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NBS

PUBLICATIONS

Proceedings

of the

Eighteenth

Joint

Meeting

U.S. DEPARTMENT OF COMMERCE

National Bureau

of Standards

•QC' 100

.056

87-3540

1987
nrB he National Bureau of Standards' was established by an act of Congress on March 3, 1901. The M Bureau’s overall goal is to strengthen and advance the nation’s science and technology and facilitate their effective application for public benefit. To this end, the Bureau conducts research and provides: (1) a

basis for the nation’s physical measurement system, (2) scientific and technological services for industry and

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The National Measurement Laboratory

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Conducts research and provides measurements, data, standards, reference

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'Headquaricrs and Laboratories at Gaithersburg, MD, unless otherwise noted; mailing address Gaithersburg, MD 20899. ^Some divisions within the center are located at Boulder, CO 80303. 'LrKated at Boulder, CO, with some elements at Gaithersburg, MD.
Research Information Center NaUonal Bureau of Standards MaryJand 20899

/O ' WIND AND . SEISMIC -

EFFECTS

Gaithersburg,

PROCEEDINGS OF

THE 18th JOINT

MEETING OF

THE U.S. - JAPAN

COOPERATIVE PROGRAM

IN NATURAL RESOURCES

PANEL ON WIND AND

SEISMIC EFFEaS

Issued April 1987

Noel J. Raufaste, EDITOR

Center for Building Technology

National Engineering Laboratory

National Bureau of Standards

Gaithersburg, MD 20899

US. DEPARTMENT OF COMMERCE,

Malcolm Baldrige, Secretary National Bureau of Standards,

Ernest Ambler, Director
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PREFACE

The U.S. -Japan Cooperative Program in Natural Resources (TUNR) was established

in 1964 for the exchange of technical information and experience mutually

beneficial to the economics and welfare of both countries. The Panel on Wind

and Seismic Effects held its first joint meeting in 1969 in Tokyo, Japan. The joint meeting has been held annually ever since, alternately in Japan and the United States. The Eighteenth Joint Meeting was held at the National Bureau of Standards, Gaithersburg, Maryland, May 12-15, 1986. Under five themes, thirty-two

technical papers were presented and discussed; thirty-eight manuscripts were

authored. Ten task committees on the Panel also held their meetings during

this period. These proceedings include the program of the Eighteenth Joint Meeting, the formal panel resolutions, all technical papers, and the task committee reports

The text of the papers had been edited for clarity.

Preparation of the proceedings was partially supported by funds from the National Science Foundation, Department of State, Bureau of Reclamation, Department of Navy, Department of Army, Federal Emergency Management Administration, Nuclear Regulatory Commission, and the National Bureau of Standards

Noel J. Raufaste, Secretary

U.S. Panel on Wind and Seismic Effects
ABSTRACT

The Eighteenth Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects was held at the National Bureau of Standards, Gaithersburg,

Maryland, from May 12-15, 1986. This publication, the proceedings of the Joint Meeting, includes the program, list of members, formal resolutions,

task committee reports, and technical papers. There were five themes under which papers were presented to the Panel. These were: Theme I - WIND ENGINEERING

Theme II - EARTHQUAKE ENGINEERING

Theme III - STORM SURGE AND TSUNAMIS

Theme IV - U.S.- JAPAN COOPERATIVE RESEARCH PROGRAM

Theme V - MEXICO CITY AND CHILIAN EARTHQUAKES

Subjects covered in the papers presented include:

(1) characteristics of stongwinds;

(2) wind loads on structures and design criteria;

(3) earthquake ground motions;

(4) soil liquefaction studies and methods to improve liquefaction

resistance;

(5) seismic loads on structures and design criteria;

(6) stress analyses of pipelines during earthquakes;

(7) full-scale seismic experiments;

(8) earthquake hazard reduction program;

(9) use of the microcomputer for earthquake studies;

(10) quantitative evaluation of damages caused by winds and earthquakes;

and

,

(11) tsunami research projects.

KEYWORDS: Accelerograph; codes; design criteria; disaster; earthquakes; earthquake hazards; geotechnical engineering; ground failures; liquefaction; pipeline; seismicity; solids; standards; structural engineering; structural response; tsunami; wind loads; and winds.

iv
CONTENTS

PAGE PREFACE iii ABSTRACT iv CONTENTS V PROGRAM FOR 18th JOINT UJNR MEETING lx

LIST OF MEMBERS xv

LIST OF TASK COMMIHEE MEMBERS xiv

LIST OF JAPANESE PARTICIPANTS xvi FORMAL RESOLUTIONS xvii THEMES AND TECHNICAL PAPERS 1 THEME I: Wind Engineering The Effects of Large-Scale Upstream Gusfing on the

Aeroelastic Behavior of Section Models 3 D.R. Huston, H.R. Bosch, and H. Tanaka On the Evaluation of Wind-Inducted Vibrations of Long-Span

Bridges Using Taut-Strip Models 18 Nobuyuki Narita, Hirohiko Tada, Koichi Yokoyama, Hiroshi Sato, Kazuo Kanzaki , and Makoto Fukuda

The Response of Taut Strip Models; Comparisons of Theoretical and Experimental Results 32 R.H. Scanlan and R.D. Marshall A Full-Scale Measurement of Wind Load on a Solar Water Heater 46 Hisashi Okada and Tatsuo Murota

Design Criteria for Glass Cladding Subjected to Wind Loads 54 Emil Simiu and Erik M. Hendrickson THEME II: Earthquake Engineering

Spectral Characteristics of Ground Motion in the Tokyo Metropolitan Area 71 Keiichi Ohtani and Shigeo Kinoshita

Strong-Motion Recordings from the Hollister, California, Area M5.5 Earthquake 26 January 1986 78

A. Gerald Brady, Richard P. Maley, and G. Noel Bycroft

V
CONTENTS

PAGE Dens© Instrument Array System by PWRI for Observing Strong Earthquake Motion o. = <. . 91 Toshio Iwasaki , Yasushi Sasaki, Keiichi Tamura, Koh Aizawa and Kazuyuki Takahashi

Vibrating Tests of an Asphaltic Facing Cofferdam

by Blastings Ill

N. Matsumoto, N. Yasuda, and M. Shiga

Summary of the Richard B. Russell Concrete Dam

Vibration Study 128 Vincent P. Chiarito and Robert J. Smith

Shaking Table Tests on Seismic Stability of Reinforced Embankments . 141 Yasuyuki Koga, Eiichi Taniguchi , and Yoshihiro Ito

Status Report on New Study Of Lower Van Norman Dam ....... 151 A.G. Franklin

Seismic Design Procedure and Dynamic Strength of CAB ...... 155 Toshio Iwasaki, Kazuhiko Kawashima, Naomi Obinata,

and Shigeki Unjoh

Large-Scale Bridge Column Tests - A Progress Report 173 Edgar V. Leyendecker, William C. Stone, and Geraldine S. Cheok

Outline of Manual for Repair Methods for Civil Engineering Structures Damaged by Earthquakes .......... 182 Toshio Iwasaki, Shigetoshi Kobayashi, Takashi Uda, Yohtaroh Asai , Yasuyuki Koga, Yasushi Sasaki, and Kazuhiko Kawashima

Effects of Vortical Construction Joints on the Dynamic

Response of Arch Dams 222 H.L. Boggs and J.R. Mays

Decision Analysis for Embankment Dams Under Seismic Load 234 T.A. Luebke A New Criterion for Assessing Liquefaction Potential 248 Susumu lai, Hajime Tsuchida, and Setsuo Noda

Earthquok© Safety Investigation of Existing Facilities ..... 257 Joseph V. Tyrrell

Thematic Mapping Works for Seismic Disaster Prevention

in Geographical Surrey Institute, Japan 263 Masatoshi Nagaoka and Hiroyuki Matsuda
CONTENTS

1

]

]

]

]

]

]

3

1

J

3

3

1

1

J1

-J

3

]

THEME III: Storm Surge and Tsunami

Satellite Methods and Applications in Marine Environmental

AnalysisC.S. Barrientos, F.G. Everdale and R.P. Stumpf

Discrimination of Flooded Areas Caused by Tsunami or Storm Surge Using Satellite Data Atsushi Takeda, Yoshlnobu Tsuji, and Hiroshi Takahashi

Fault Displacement Seiche Waves on Inland Reservoir and Lakes . „ . , , Robert L. Dewey and Karl M. Dise THEME IV: U.S.-Japan Cooperative Research Program

U.S.- Japan Coordinated Earthquake Research Program on

Reinforced Masonry Buildings—Seismic Capacity of Reinforced Masonry Walls and BeamsShin Okamoto, Yutaka Yamazaki, Takashi Kaminosono,

Masaorai Teshigawara, and Hisahiro Hiraishi

U.S.- Japan Coordinated Program on Masonry Research Shin Okamoto and James Noland

Seismic Behavior of a Full-Scale K-Brace Steel Building

StructureHiroyuki Yamanouchi, Mitsumasa Mldorikawa, Isao Nishiyama,

and Tatsuo Murota

Full-Scale Studies of Steel Structures Robert D. Hanson THEME V: Mexico City and Chilian Earthquakes

Earthquake Damage in Mexico Caused by the September 1985

Earthquakes Edgar V. Leyendecker, William C. Stone, Felix Y. Yokel

Japan-Mexico Technical Cooperation on Damage Evaluation, Repair and Strengthening for Buildings Damaged by 1985 Mexico Earthquake Masaya Hirosawa and Tatsuo Murota

Aspects of Strong Motion from the Michoacan, Mexico, Earthquake

of September 19, 1985 John G. Anderson, James N. Brune, Jorge Prince, Enrique Mena, Paul Bodin, Mario Onate, Roberto Ouaas

,

and Shri Krishna Singh

Seismic Damage to Port Facilities by the 1985 Chile Earthquake Improvement of Seismic Safety of the Damaged Ports Hajirae Tsuchida and Setsuo Noda

PAGE

275

285

289

307

320

328

344

357

368

372

391
CONTENTS

PAGE

Observation of the Structural Damage Caused by the 3 March 1985

Chile Earthquake ....... ..... . 402 Mete A. Sozen

Papers Not Presented Orally But Included In The Proceedings

Spatial Structure of the Wind in the Marine Boundary Layer . . 409 Keikichi Naito, Tatsuo Hanafusa, and Tokunosuke Fujitani

Loading Test on the Connection of Concrete Filled Steel Tubes „ , 4 1

3

Hirohiko Tada, Hiroshi Shinohara, Koichi Minosaku, Akira Takizawa, and Jun Hikino

Deformation Capacity of Steel Truss Girder 429 Isao Nishiyama, Tatsuo Murota, Nagahide Kani, and Shin-ichi Ina

Field Investigation of Seismically Damaged Sites by Use of Vibratory Cone Penetrometer 443 Yasuyuki Koga, Yasushi Sasaki, Yoshihiro Itoh, and Osamu Matsuo An Investigation on a Twelve-Story Office Building Damaged by the 1985 Mexico Earthquake 454 Hisahiro Hiraishi, Masaya Hirosawa, Hideo Tsukagoshi, Toshihide Kashima, and Toshikazu Kawashima

Practical Approach for Determining Stage Frequency 468

H. Lee Butler and Thomas A. Hardy

Appendix: Ten Task Committee Reports 473
PROGRAM FOR 18th JOINT UJNR MEETING

SATURDAY 10 MAY

1800, DCA NW-4 JAPANESE TEAM ARRIVES WASHINGTON DC

2000, INFORMAL GATHERING HOSTED BY US SIDE IN GAITHERSBURG

HOTEL, HOLIDAY INN GAITHERSBURG

SUNDAY 11 MAY

1830, RECEPTION HOSTED BY DR. AND MRS WRIGHT AT THEIR

HOME MONDAY 12 MAY OPENING CEREMONIES (lOOO - 1145) 1000 CALL TO ORDER BY NOEL RAUFASTE, SECRETARY

US SIDE UJNR PANEL OPENING REMARKS BY DR. ERNEST AMBLER, DIRECTOR NATIONAL BUREAU OF STANDARDS

REMARKS BY MR. KANAME IKEDA, COUNSELLOR FOR SCIENCE AND TECHNOLOGY, EMBASSY OF JAPAN

REMARKS BY DR. RICHARD WRIGHT, CHAIRMAN

US SIDE UJNR PANEL REMARKS BY DR. MASATERU TOMINAGA, CHAIRMAN JAPANESE SIDE INTRODUCTION OF US MEMBERS BY US CHAIRMAN

INTRODUCTION OF JAPANESE MEMBERS BY JAPANESE CHAIRMAN

ELECTION OF TECHNICAL SESSION CHAIRMEN

APPROVAL OF AGENDA

GROUP PHOTOGRAPH

1145 LUNCH (HOSTED BY DR. ERNEST AMBLER) TECHNICAL SESSIONS

SESSION 1 - WIND ENGINEERING

CHAIRMAN R.N. WRIGHT

1300 - 1320 THE EFFECTS OF LARGE-SCALE UPSTREAM

GUSTING ON THE AEROELASTIC BEHAVIOR OF SECTION MODELS, BY D. HUSTON, H. BOSCH, H. TANAKA (Presented by D. HUSTON)

1320-1340 ON THE EVALUATION OF WIND- INDUCED VIBRATIONS OF LONG-SPAN BRIDGES USING TAUT-STRIP

MODELS, BY N. NARITA, H. TADA, K. YOKOYAMA, H. SATO,

K. KANZAKI, M. FUKUDA (Presented by H. TADA)

1340 - 1400 THE RESPONSE OF TAUT STRIP MODELS, COMPARISONS OF THEORETICAL AND EXPERIMENTAL RESULTS, BY R. SCANLAN, R. MARSHALL (Presented by R. MARSHALL)

1400 - 1420 DISCUSSIONS 1420-1440 A FULL-SCALE MEASUREMENT OF WIND LOAD ON A SOLAR WATER HEATER, BY H. OKADA, T. MUROTA

(Presented by H. OKADA)

1440 - 1500 DESIGN CRITERIA FOR GLASS SUBJECTED TO WIND LOADS, BY E. SIMIU, E. HENDRICKSON (Presented

by E. SIMIU)

1500 - 1520 DISCUSSION

1520 - 1540 BREAK

ix
SESSION 2 - EARTHQUAKE ENGINEERING

CHAIRMAN R.N. WRIGHT

1540 - 1600 MICROZONING AND PROBABILISTIC HAZARD

ASSESSMENT, BY T. ALGERMISSEN

1600 - 1620 SPECTRAL CHARACTERISTICS OF GROUND MOTION

IN THE TOKYO METROPOLITAN AREA, BY K. OHTANI

,

S. KINOSHITA (Presented by K. OHTANI)

1620 - 1640 STRONG-MOTION RECORDINGS FROM THE

HOLLISTER, CALIFORNIA, AREA M5 . 5 EARTHQUAKE, JANUARY 26, 1986, BY A.G. BRADY, RoP. MALEY,

GoN. BYCROFT (Presented by A. Go BRADY)

1640 - 1700 DENSE INSTRUMENT ARRAY SYSTEM BY PWRI FOR OBSERVING STRONG EARTHQUAKE MOTION, BY T. IWASAKI

,

Y. SASAKI, K. TAMURA, K. AIZAWA, K. TAKAHASHI

(Presented by Y. SASAKI)

1700 - 1720 DISCUSSION

1720 ADJOURN

1930 DINNER - BULL FEATHERS RESTAURANT OF CAPITOL HILL

(HOSTED BY US MEMBERS) TUESDAY 13 MAY

SESSION 3 - EARTHQUAKE ENGINEERING

CHAIRMAN M. TOMINAGA

840 - 900 VIBRATION TESTS OF AN ASPHALTIC FACING COFFERDAM

BY BLASTINGS, BY N. MATSUMOTO, N. YASUDA,

M. SHIGA (Presented by H. TADA)

900 - 920 SUMMARY OF THE RICHARD B. RUSSELL CONCRETE DAM VIBRATION STUDY, BY V.P. CHIARITO, R.J. SMITH (Presented by R.J. SMITH)

920 - 940 SHAKING TABLE TEST ON SEISMIC STABILITY OF REINFORCED EMBANKMENTS, BY Y. KOGA, E. TANIGUCHI

,

Y. ITO (Presented by Y. SASAKI)

940 - 1000 DISCUSSION

1000 - 1020 STATUS REPORT ON NEW STUDY OF LOWER VAN NORMAN DAM, BY A.G. FRANKLIN

1020 - 1040 SEISMIC DESIGN PROCEDURE AND DYNAMIC STRENGTH OF "CAB", BY T. IWASAKI, K. KAWASHIMA,

N. OBINATA, S. UNJOH (Presented by H. TADA)

1040 - 1100 LARGE-SCALE BRIDGE COLUMN TESTS - A PROGRESS REPORT, BY E.V. LEYENDECKER, W. STONE, G. CHEOCK (Presented by E.V. LEYENDECKER)

1100 - 1120 DISCUSSION

1120 - 1140 BREAK

1140 - 1200 OUTLINE OF MANUAL FOR REPAIR METHODS FOR CIVIL ENGINEERING STRUCTURES DAMAGED BY EARTHQUAKES,

BY T. IWASAKI, S. KOBAYASHI , T. UDA, Y. ASAI

,

Y. KOGA, Y. SASAKI, K. KAWASHIMA (Presented by

Y. SASAKI)

1200 - 1220 EFFECTS OF VERTICAL CONTRACTION JOINTS ON THE DYNAMIC RESPONSE OF ARCH DAMS, BY H. BOGGS,

J. MAYS (Presented by H. BOGGS)

1220 - 1240 DISCUSSION

1240 LUNCH (HOSTED BY DR. EDWARD BRADY, ASSOCIATE DIRECTOR FOR INTERNATIONAL AFFAIRS)

X
SESSION 4 EARTHQUAKE ENGINEERING

CHAIRMAN R.N. WRIGHT

1340 - 1400 DECISION ANALYSIS FOR EMBANKMENT DAMS UNDER SEISMIC LOADS, BY T. LUEBKE (Presented by

F. McLEAN)

1400-1420 A NEW CRITERION FOR ASSESSING LIQUEFACTION

POTENTIAL, BY S. lAI , H. TSUCHIDA, S. NODA (Presented

by H. TSUCHIDA)

1420 - 1440 EARTHQUAKE SAFETY INVESTIGATION OF EXISTING FACILITIES BY J. TYRRELL

1440 - 1500 THEMATIC MAPPING WORKS FOR SEISMIC DISASTER PREVENTION IN GEOGRAPHICAL SURVEY INSTITUTE,

JAPAN, BY M. NAGAOKA, H. MATSUDA, (Presented by

M. NAGAOKA)

1500 - 1520 DISCUSSION

1520 - 1540 BREAK TASK COMMITTEE MEETINGS (1540 - 1800) TASK COMMITTEE A, STRONG MOTION INSTRUMENTATION

ARRAYS AND DATA TASK COMMITTEE B, LARGE SCALE TESTING PROGRAM

TASK COMMITTEE D, EVALUATION OF PERFORMANCE OF STRUCTURES TASK COMMITTEE H, SOIL BEHAVIOR AND STABILITY

DURING EARTHQUAKES TASK COMMITTEE J, WIND AND EARTHQUAKE ENGINEERING

FOR TRANSPORTATION SYSTEMS

1900 DINNER - JAPANESE MEMBERS HOSTED BY US MEMBERS

WEDNESDAY 14 MAY

SESSION 5 STORM SURGE AND TSUNAMI CHAIRMAN M. TOMINAGA

900 - 920 SATELLITE APPLICATION IN MARINE AREAS, BY

C. BARRIENTOS

920 - 940 DISCRIMINATION OF FLOODED AREAS CAUSED BY TSUNAMI OR STORM SURGE USING SATELLITE DATA, BY

A. TAKEDA, Y. TSUJI , H. TAKAHASHI (Presented by

K. OHTANI)

940 - 1000 FAULT DISPLACEMENT SEICHE WAVES ON INLAND RESERVOIR AND LAKES, BY R. DEWEY, K. DISE (Presented by R. DEWEY)

1000 - 1020 DISCUSSION

1020 - 1040 BREAK
SESSION 6 US JAPAN COOPERATIVE RESEARCH PROGRAM CHAIRMAN M. TOMINAGA

1040 - 1100 US-JAPAN COORDINATED EARTHQUAKE RESEARCH PROGRAM ON REINFORCED MASONRY BUILDINGS, BY

S. OKAMOTO, Y. YAMAZAKI , T. KAMINOSONO,

Mo TESHIGAWARA, H. HIRAISHI (Presented by

Ho YAMANOUCHI)

1100 - 1120 US JAPAN COORDINATED PROGRAM ON MASONRY

RESEARCH, BY S. OKAMOTO, J. NOLAND (Presented by

M. PORTER)

1120 - 1140 SEISMIC BEHAVIOR OF A FULL-SCALE K-BRACED

STEEL BUILDING STRUCTURE, BY YAMANOUCHI,

M. MIDORIKAWA, I. NISHIYAMA, T. MUROTA (Presented

by H. YAMANOUCHI)

1140 - 1200 FULL-SCALE STUDIES OF STEEL STRUCTURES, BY R. HANSON

1200 - 1220 DISCUSSION

1220 - LUNCH (HOSTED BY DR. JOHN LYONS, DIRECTOR, NATIONAL ENGINEERING LABORATORY) TASK COMMITTEE MEETINGS (1400 - 1700) TASK COMMITTEE C, REPAIR AND RETROFIT OF EXISTING

STRUCTURES TASK COMMITTEE E, LAND USE PROGRAM FOR CONTROLLING NATURAL HAZARD EFFECTS TASK COMMITTEE F, DISASTER PREVENTION METHODS FOR LIFELINE SYSTEMS TASK COMMITTEE G, WIND CHARACTERISTICS AND STRUCTURAL RESPONSE TASK COMMITTEE I, STORM SURGE AND TSUNAMIS

1800 DINNER JAPANESE MEMBERS HOSTED BY US MEMBERS THURSDAY 15 MAY

SESSION 7 MEXICO CITY AND CHILIAN EARTHQUAKES, CHAIRMAN M. TOMINAGA

900 - 920 EARTHQUAKE DAMAGE IN MEXICO CAUSED BY THE SEPTEMBER 1985 EARTHQUAKES, BY E.V. LEYENDECKER,

W. STONE, F. YOKEL (Presented by E.V. LEYENDECKER)

920 - 940 JAPAN-MEXICO TECHNICAL COOPERATION ON DAMAGE EVALUATION, REPAIR AND STRENGTHENING FOR BUILDINGS DAMAGED BY 1985 MEXICO EARTHQUAKE, BY

M. HIROSAWA, T. MUROTA (Presented by H. OKADA)

940 - 1000 ASPECTS OF STRONG MOTION FROM THE MICHOACAN, MEXICO EARTHQUAKE OF SEPTEMBER 19, 1985, BY J. ANDERSON, J. BRUNE, J. PRINCE, E. MENA, P. BODIN,

M. ONATE, R. QUAAS, S.K. SINGH (Presented by J. ANDERSON

)

1000 - 1020 DISCUSSION

1020 - 1040 BREAK

1040 - 1100 SEISMIC DAMAGE TO PORT FACILITIES BY THE

1985 CHILE EARTHQUAKE AND IMPROVEMENT OF SEISMIC SAFETY OF THE DAMAGED PORTS, BY H. TSUCHIDA,

S. NODA (Presented by H. TSUCHIDA)

1100 - 1120 OBSERVATIONS OF THE STRUCTURAL DAMAGE CAUSED BY THE 3 MARCH 1985 CHILE EARTHQUAKE, BY

M. SOZEN

1120 - 1140 DISCUSSION

1140 - 1200 BREAK

1200 - 1300 LUNCH (HOSTED BY DR. RICHARD WRIGHT)

xii
SESSION 8 REPORT OF TASK COMMITTEES CHAIRMAN R.N. WRIGHT

1300

1440

1500

1530

1830

1440A STRONG MOTION INSTRUMENTATION ARRAYS AND DATA

B LARGE SCALE TESTING PROGRAM

C REPAIR AND RETROFIT OF EXISTING STRUCTURES

D EVALUATION OF PERFORMANCE OF STRUCTURES

E LAND USE PROGRAM FOR CONTROLLING NATURAL HAZARD EFFECTS

F DISASTER PREVENTION METHODS FOR LIFELINE

SYSTEMS

G WIND CHARACTERISTICS AND STRUCTURAL RESPONSE

H SOIL BEHAVIOR AND STABILITY DURING EARTHQUAKES

I STORM SURGE AND TSUNAMIS

J WIND AND EARTHQUAKE ENGINEERING FOR TRANSPORTATION SYSTEMS

1500 BREAK

1530 ADOPTION OF FINAL RESOLUTIONS AND ADOPTION OF CHARTER FOR UJNR PANEL ON WIND

AND SEISMIC EFFECTS

1600 CLOSING SESSION

CALL TO ORDER BY N.J. RAUFASTE

CLOSING REMARKS BY M. TOMINAGA

CLOSING REMARKS BY R.N. WRIGHT

ADJOURN DINNER HOSTED BY JAPANESE MEMBERS

(HISAGO RESTAURANT, WASHINGTON HARBOR)
STUDY TOUR 16-24 MAY

FRIDAY 16 MAY

STUDY TOUR HOSTED BY US SIDE TO THE BALTIMORE HARBOR TUNNEL

AND BALTIMORE HARBOR RECONSTRUCT!ON/RENOVATION

1830 DINNER HOSTED BY US MEMBERS SATURDAY 17 MAY VISIT BRIDGE SITES AND THE WORLD TRADE CENTER (Hosted

by ASCE)

900 EA-1420 LV DCA

1000 AR NYC

HOTEL; DORAL HOTEL SUNDAY 18 MAY

TEAM TRAVELS TO VICKSBURG, MS

1101 EA-543 LV LGA

1427 EA-595 AR JACKSON, MS

1600 ARRIVE VICKSBURG

HOTEL; RAMADA INN, Vicksburg MONDAY 19 MAY

900 TEAM VISIT TO WATERWAYS EXPERIMENT STATION

1400 DEPART BY AUTO TO JACKSON

1635 DP JACKSON DL-557

2010 AR SAN FRANCISCO DL-143

HOTEL; BEVERLY PLAZA HOTEL TUESDAY 20 MAY

1000 TEAM VISIT TO USGS MENLO PARK WEDNESDAY 21 MAY

TEAM TRAVELS TO HONOLULU

930 NW-101 LV SAN FRANCISCO

1135 ARRIVES HONOLULU VISIT PACIFIC TSUNAMI WARNING CENTER AT EWA BEACH

91-270 Ft. Weaver Road, Ewa Beach

1915 LV HONOLULU AQ-48

1955 AR HILO

HOTEL; HILO HAWAIIAN THURSDAY 22 MAY

900 TEAM VISITS HAWAIIAN VOLCANO OBSERVATORY

1535 LV HILO

1615 AR HONOLULU

HOTEL; KAIMANA BEACH HOTEL,

FRIDAY 23 MAY TEAM DEPARTS FOR JAPAN

1105 NW-009

xiv
LIST OF MEMBERS

U.S. PANEL ON WIND AND SEISMIC EFFECTS MEMBERSHIP LIST

1986

Dr. Richard N. Wright, Chairman Dr. John J. Burns, Jr. Director, Center for Building Technology Head, Structures Section National Bureau of Standards Office of Research

Gaithersburg, MD 20899 U.S. Nuclear Regulatory Commission

(301) 921-3377 Washington, DC 20555

(301) 443-7860

Mr. Noel J. Raufaste, Secretary US Side Structures Division

Center for Building Technology

National Bureau of Standards Gaithersburg, MD 20899

(301) 921-3106

Dr. S. T. Algermlssen

Office of Earthquake Studies Branch of Earthquake Tactonlcs, USGS

Denver Federal Center

Stop 978, Box 25046 Denver , CO 80225

(303) 236-1611

Dr. Celso S. Barrientos Supervisory Physicist Scientist NOAA/NESDIS

3300 vniltehaven Street, NW

Washington, DC 20235

(202) 634-7379

Dr. Eddie N. Bernard

Director, Pacific Marine

Environmental Laboratory

7600 San Point Way, NE

Seattle, WA 98115

(206) 526-6239 FTS 292-6800

Dr. Roger D. Borcherdt

Chief, Branch of Engineering

Seismology and Geology

Office of Earthquake Studies, USGS

345 Mlddlefleld Road Menlo Park, CA 94025

(415) 323-8111 X2755

Dr. A. Gerald Brady

Physical Scientist Office of Earthquake Studies

U.S. Geological Survey

345 Mlddlefleld Road Menlo Park, CA 94025

(415) 323-8111

Mr. H. Lee Butler

Chief, Research Division

Coastal Engineering Research Center Waterways Experiment Station

U.S. Army Corps of Engineers

Vicksburg, MS 39108

(601) 634-3462

Dr. James Choromokos, Jr. Director, Research and Development

Department of the Army DAEN-RDZ-A

Washington, DC 20314

(202) 272-0254

Dr. A. G. Franklin

Chief, Earthquake Engineering

and Geophysics Division

Geotechnical Laboratory

U.S. Army Engineer Waterways Experiment Station Vicksburg, MS 39180

(601) 634-2658

Mr. G. Robert Fuller

Chief, Standards Branch

Manufactured Housing Construction

Standards Division, Room 9156 Housing and Urban Development

Washington, DC 20410-8000

(202) 755-6920

Mr. James Gates

Office of Structures Design

Department of Transportation

1120 N Street Sacramento, CA 95807

(916) 445-1439

XV
Dr. Michael P. Gaus Section Head of Critical Engineering Division of Emerging and Critical Engineering Systems National Science Foundation

1800 G Street, NW

Washington, DC 20550

(202) 357-9500

Mr. Peter E. Gurvln

A/FBO, SA-6 Department of State Washington, DC 20620

(202) 235-3689

Dr. Walter W. Hays Deputy for Research Application

n.S. Geological Survey

Office of Earthquake Survey

905 National Center

Reston, VA 22043

FTS 928-6472 or (703) 860-6471

Dr. William B. Joyner Geophysicist

Office of Earthquake Studies Branch of Ground Motion and Faulting, nSGS

342 Mlddlefield Road Menlo Park, CA 94025

(415) 323-8111

Mr. Richard W, Kramer

Civil Engineer

Technical Review Staff, D-3210

Bureau of Reclamation

P.O. Box 25007 Denver, CO 80225

FTS 776-8539 or (303) 236-8539

Mr. Richard Rrlmm

Office of Natural and Technological Hazards

Federal Emergency Management Agency

Washington, DC 20472

(202) 646-2871

Mr, Paul K, Krumpe

Office of Foreign Disaster Assistance Agency for International Development Department of State Washington, DC 20523

(202) 647-9758

Mr. James F. Lander Deputy Director

National Geophysical Data Center NOAA/NESDIS

325 Broadway

Boulder, CO 80303

FTS 320-6215 or (303) 497-6215

Dr. H. S. Lew

Leader, Construction Safety Group

Structures Division

Center for Building Technology

National Bureau of Standards Gaithersburg, MD 20899

(301) 921-2647

Dr. E. V. Leyendecker

Leader, Earthquake Hazards Reduction

Group, Structures Division

Center for Building Technology

National Bureau of Standards Gaithersburg, MD 20899

(301) 921-3471

Dr. Shlh-Chl Liu

Director, Earthquake Hazards Mitigation

National Science Foundation

1800 G Street, N.W. Room 1128 Washington, DC 20550

(202) 357-9780

Mr. Robert MacDonald

Geological Services Branch, Code D

1630 Bureau of Reclamation

Denver Federal Center, Building 67 Denver, CO 80225

FTS 776-4195 or (303) 236-4915

Dr> Richard Marshall Structural Research Engineer

Structures Division

Center for Building Technology

National Bureau of Standards Gaithersburg, MD 20899

(301) 921-2170

Dr. Richard D. McConnell

Office of Construction Veterans Administration

811 Vermont Avenue, NW

Washington, DC 20420

(202) 389-3103

xvi
Dr. Francis G. McLean

Chief, Division of Research and Laboratory Services, D-1500

Bureau of Reclamation

P.O. Box 25007 Denver, CO 80225

FTS 776-5981 or (303) 234-3303

Mr. John F. Meehan

Research Director

Structural Safety Section

Office of the Architect

Department of General Services

State of California

Sacramento, CA 95805

(916) 445-8730

Dr. John B. Scalzi Program Director, Systems Engineering for Large Structures National Science Foundation

1800 G Street, NW, Room 1130 Washington, DC 20550

(202) 357-7710

Mr. Joseph T. Schaefer

Chief, Scientific Services Division

National Weather Service, NOAA

Federal Building, Room 1836

601 East 12th Street Kansas City, MO 64106

(816) 374-5672

Dr. Charles Scheffey

Science Advisor, Offices of Research, Development and Technology

Federal Highway Administration

6300 Georgetown Pike

McLean, VA 22101

(703) 285-2054

Mr. Robert J. Smith Office of the Chief of Engineers

Department of the Army DAEN-ECE-D

Washington, DC 20314-1000

(202) 272-0220

Dr. Lawrence A. Soltis Supervisory Research Engineer

Engineering Design Criteria

Forest Products Laboratory

Forest Service

U.S. Department of Agriculture

Madison, WI 53705

(608) 364-5910

Mr. Joseph Tyrrell

Director, Civil/Structural Div. (04BA) Naval Facilities Engineering Command

200 Stovall Street Alexandria, VA 22332

(703) 325-0047 ALTERNATES

Dr. Clifford Astill

National Science Foundation

1800 G Street, NW

Washington, DC 20550

(202) 357-9500

Mr. Michael Changery

National Climatic Center

Federal Building

Asheville, NC 28787

(704) 259-0765

Mr. Vincent Chiarito

U.S. Army

Waterways Experiment Station

P.O. Box 631 Vicksburg, MS 39180

FTS 542-2000

Dr. A. J. Eggenberger

National Science Foundation

1800 G Street, NW

Washington, DC 20550

(202) 357-9500

Mr. Lucian G. Guthrie

Structural Engineer

Office Chief of Engineering

Engineering Division

Directorate of Engineering

and Construction

HQUSACE (DAEN-ECE-D) Washington, DC 20314-1000

(202) 272-8673

Mr. James Houston

Division Chief Waterways Experiment Station

P.O. Box 631 Vicksburg, MS 39180-0631

FTS 542-2000
Mr. Roger Kenneally

Structures Division

Office of Research

U.S. Nuclear Regulatory Conanission Washington, DC 20555

(301) 443-7860

Mr. Robert R. Ledzian

Senior Staff Assistant for Research

Bureau of Reclamation, WO-220

Department of the Interior 18th and C Streets, NW

Washington, DC 20240

(202) 343-6703

Mr. T. K. Lew

Commanding Officer Naval Civil Engineering Laboratory

Port Hueneme, CA 93043

(805) 982-5785

Dr. Herbert Meyers

Chief, Solid Earth Geophysics Division

National Geophysical Data Center NOAA

325 Broadway

Boulder, CO 80303

(303) 497-6215

Ms. Janina Z. Mlrski

Chief, Southem/Westem Structural Division 0853 Office of Construction

Veterans Administration

Washington, DC 20420

(202) 389-2394

Mr. Ronald J. Morony

Program Manager

Building Technology Research Staff Department of Housing and Urban Development

7th and D Street, SW

Washington, DC 20410

(202) 755-0640 TEMPORARY

Mr. Harold H. Boggs Bureau Reclamation

P.O. Box 25007 Denver, CO 80225

(303) 234-3303

Mr. Harold R. Bosch

Research Structural Engineer

Structures Division HNR-10

Federal Highway Administration

6300 Georgetown Pike

McLean, VA 22101

(703) 285-2446

Mr. Robert L. Dewey

Civil Engineer Bureau of Reclamation

P.O. Box 25007

Denver, CO 80225

(303) 234-3303

Mr. Karl M. Dise

Civil Engineer

Bureau of Reclamation

P.O. Box 25007

Denver, CO 80225

(303) 234-3303 Professor Robert D. Hanson

Department of Civil Engineering

University of Michigan

2340 GG Brown Building

Ann Arbor, MI 48109-2125

(313) 764-5617

Mr. T. A. Luebke Bureau of Reclamation

P.O. Box 25007 Denver, CO 80225

(303) 234-3303

Mr. J. R. Mays

University of Colorado

Denver, CO 80202

Mr. James Noland

Principal, Atklnson-Noland Associates Consulting Engineers

2619 Spruce Street

Boulder, CO 80302

(303) 444-3620

Professor Max Porter

Department of Civil Engineering

Iowa State University

416 A Town Engineering Building

Ames , Iowa 50011

(515) 294-7456

Professor Robert Scanlan Department of Civil Engineering

Johns Hopkins University

202 Lathrobe Hall

Baltimore, MD 21218

(301) 338-7138

Mr. Emil Simlu

Structural Research Engineer

Structures Division

Center for Building Technology

National Bureau of Standards Gaithersburg, MD 20899

(301) 921-3169
JAPAN PANEL ON WIND AND SEISMIC EFFECTS MEMBERSHIP LIST

1986

Dr. Masateru Tominaga, Chairman

Director-General

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Toshiaki Fujii, Secretary-General

Director Planning and Research Administration Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi,

Tsukuba-gun , Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Yoshimi Goda

Deputy Director-General

Port and Harbour Research Institute Ministry of Transport

3-1-1, Nagase, Yokosuka-shi , Kanagawa-ken 239

Tel. 0468-41-5410

Dr. Tatsuo Hanafusa

Chief, Second Research Laboratory

Physical Meteorology Research Division

Meteorological Research Institute (MRI) Japan Meteorological Agency (JMA)

1-1, Nagamine, Yatabe-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-51-7111

Dr. Sadaiku Hattori

Director

International Institute of Seismology and Earthquake Engineering (I.I.S.E.E.) Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Masaya Hirosawa

Director

Research Planning and Information Department Building Research Institute Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Masaji Ichikawa

Head, Seismology and Volcanology Division

Meteorological Research Institute (MRI) Japan Meteorological Agency (JMA)

1-1, Nagamine, Yatabe-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-51-7111

xix
Dr. Yuji Ishiyama Head of Planning and Investigation Division

Research Planning and Information Department Building Research Institute Ministry of Construction

Tatehara 1-banchi , Oh-ho-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Toshio Iwasaki Director, Earthquake Disaster Prevention Department

Public Works Research Institute Ministry of Construction

Asahi 1-banchi, Toyosato-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Kazuhiko Kawashima

Head, Earthquake Engineering Division

Earthquake Disaster Prevention Department

Public Works Research Institute Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Yasuyuki Koga

Head, Soil Dynamics Division Construction Method and Equipment Department

Public Works Research Institute Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Keizo Masamura

Head, Typhoon Research Division

Meteorological Research Institute (MRI) Japan Meteorological Agency (JMA)

1-1, Nagamine, Yatabe-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-51-7111

Dr. Norihisa Matsumoto

Head, Fill Type Dam Division

Dam Department

Public Works Research Institute Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tatsuo Murota Director

Structural Engineering Department

Building Research Institute Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

XX
Mr. Masatoshi Nagaoka

Head of Second Geographic Division

Geographical Survey Institute

Ministry of Construction

Kitazato 1-banchi , Yatabe-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-1111

Dr. Nobuyuki Narita

Deputy Director-General

Public Works Research Institute Ministry of Construction

Asahi 1-banchi, Toyosato-machi

Tsukuba-gun, Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Setsuo Noda

Chief, Earthquake Resistant Structures Laboratory Structure Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi , Kanagawa-ken 239

Tel. 0468-41-5410

Mr. Keiichi Ohtani

Head, Earthquake Engineering Laboratory

Second Research Division

National Research Center for Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Sakura-mura, Niihari-gun

Ibaraki-ken 305

Tel. 0298-51-1611

Mr. Michio Okahara

Head, Foundation Engineering Division

Structure and Bridge Department

Public Works Research Institute Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Shin Okamoto

Director, Production Department

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibraki-ken 305

Tel. 0298-64-2151

Dr. Yasushi Sasaki

Head, Ground Vibration Division

Earthquake Disaster Prevention Department

Public works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211
Mr. Hiroshi Shinohara

Head, Bridge Division

Structure and Bridge Department

Public Works Research Institute Ministry of Construction

Asahi 1-banchi, Toyosato-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Yoshihiro Sugimura

Head, Foundation Engineering Division

Structure Engineering Department

Building Research Institute Ministry of Construction Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Hirohiko Tada

Director, Structure and Bridge Department

Public Works Research Institute Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Hiroshi Takahashi Director-General il National Research Center for Disaster Prevention

Science and Technology Agency f 3-1, Tennodai , Sakura-mura, Niihari-gun 1 Ibaraki-ken 305 I Tel. 0298-51-1611

Dr. Hajime Tsuchida

Director, Structures Division

Port and Harbour Research Institute Ministry of Transport

3-1-1, Nagase, Yokosuka-shi , Kanagawa-ken 239

Tel. 0468-41-5410

Dr. Takaaki Uda

Head, Coastal Engineering Division

River Department

Public Works Research Institute Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Hiroyuki Yamanouchi

Head, Structure Division

Structural Engineering Department

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

xii
Mr. Koichi Yokoyama

Head, Structure Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

xiii
Task Committee

LIST OF TASK COMMIHEE MEMBERS

A. Strong-Motion Instrumentation Arrays & Data

B. Large-Scale Testing

Programs

C. Repair and Retrofit of Existing Structures

D. Evaluation of Performance

E. Natural Hazard Effects

Land Use Program for Controlling

F. Disaster Prevention Methods

for Lifeline Systems

G. Wind Characteristics and Structural Response

H. Soil Behavior and Stability

During Earthquakes

U.S. Side *A.G. Brady

W.B. Joyner

J.F. Lander

F.G. McLean

*E.V. Leyendecker

H.S. Lew

J.B. Scalzi

C.F. Scheffey

*JcB. Scalzi

P.E. Gurvin

H.S. Lew

R.D. McConnell

C.F. Scheffey

*G.R. Fuller

R.D. McConnell

J.F. Meehan

J.B. Scalzi

*S.T. Algermissen

A.G. Brady

G.R. Fuller

R.D. Marshall

*S.C. Liu

J.B. Scalzi

E.V. Leyendecker

*R.D. Marshall

C.S. Barrientos

M.P. Gaus

C.F. Scheffey

*A.G. Franklin

F.G. McLean

R.J. Smith

Japanese Side *Hajime Tsuchida

Sadaiku Hattori

Masaji Ichikawa

Yuji Ishiyama Keiichi Ohtani Yasushi Sasaki *Keiichi Ohtani

Yuji Ishiyama

Yasuyuki Koga

Setsuo Noda

Shin Okamoto Koichi Yokoyama

*Masaya Hirosawa

Yuji Ishiyama Toshio Iwasaki

Shin Okamoto

Hirohiko Tada *Shin Okamoto

Yuji Ishiyama Kayuya Ohshima

Hirohiko Tada

*Kazuhiko Kawashima

Sadaiku Hattori Toshio Iwasaki Masatoshi Nagaoka Yoshihiro Sugiraura *Toshio Iwasaki Kazuhiko Kawashima

Tatsuo Murota

Setsuo Noda Keiichi Ohtani Yasushi Sasaki *Keikichi Naito

Keizo Masamura

Tatsuo Murota

Nobuyuki Narita

Koichi Yokoyama

*Yasushi Sasaki Yasuyuki Koga Norihisa Matsumoto

Kayuya Ohshima Yoshihiro Sugimura

Hajime Tsuchida
]

]

1

]

]

]

]

]

]

]

j

]

]

]

]

1

]

I. storm Surge and Tsunamis

Wind and Earthquake

Engineering for Transportation Systems

*C.S. Barrientos *Yoshimi Guda

E.N. Bernard Masai i Ichikawa

M.P. Gaus Keizo Masamura

J.F. Lander Hiroshi Takahashi

Takaaki Uda

*C.F. Scheffey *Hirohiko Tada

A.G. Franklin Toshio Iwasaki

H.S. Lew Kazuhiko Kawashima

E.V. Leyendecker Nobuyuki Narita

Kayuya Ohshima

Yasushi Sasaki Hiroshi Shinohara

Koichi Yokoyama

*Chairman

XV
LIST OF JAPANiSE PARTICIPANTS

Member

Dr. Masateru Tominaga

Director General Public Works Research Institute Ministry of Construction

Mr. Masatoshi Nagaoka

Head, Second Geographic Division

Geographical Survey Institute Ministry of Construction

Mr. Keiichi Ohtani

Chief, Earthquake Engineering Laboratory

Second Research Division

National Research Center for Disaster Prevention

Science and Technology Agency

Dr. Yasushi Sasaki

Head, Ground Vibration Division

Earthquake Disaster Prevention Department

Public Works Research Institute Ministry of Construction

Dr. Hirohiko Tada

Director, Structure and Bridge Department

Public Works Research Institute Ministry of Construction

Dr, Hajime Tsuchida

Director, Structure Division

Port and Harbour Research Institute Ministry of Transport

Dr. Hiroyuki Yamanouchi

Head, Building Structure Division

Structural Engineering Department

Building Research Institute Ministry of Construction Temporary Member

Dr. Hisashi Okada

Senior Researcher, Building Aerodynamics Division

Structural Engineering Department

Building Research Institute Ministry of Construction

Mr. Akira Endoh

Researcher, New Transportation Systems Division

Road Department

Public Works Research Institute Ministry of Construction
FORMAL RESOLUTIONS

RESOLUTIONS OF THE EIGHTEENTH JOINT MEETING

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS (UJNR) National Bureau of Standards

Gaithersburg, Maryland

May 12-15, 1986 The following resolutions are hereby adopted:

1. The Eighteenth Joint Panel Meeting provided valuable exchange of technical information which was beneficial to both countries. In view

of the importance of cooperative programs on the subject of wind and seismic effects, the continuation of Joint Panel Meetings is considered

essential.

2. The following activities have been carried out since the Seventeenth

Joint Meeting:

a. In December 1985 this Panel's activities were reported at the

11th UJNR Plenary Conference.

b. Technical documents, strong motion records, research reports, earthquake damage records, and workshop proceedings were

/ exchanged.

c. Workshops were held that continued joint research and development activities on methodologies for wind and seismic disaster mitigation.

d. The Coordinated Research Program on Lcirge-Scale Testing of Masonry

Structures which is currently underway. Research is coordinated by the Joint Technical Coordinating Committee on Masonry Research (JTCCMAR).

3. A Charter was drafted for this Panel on Wind and Seismic Effects as suggested at the 11th UJNR Plenaury Conference, December 1985, Tokyo

(attached). The draft will be studied and considered for adoption at the nineteenth joint meeting.

4. The Panel endorses the following workshops and conferences proposed by the Task Committees:

a. Planning conference on Disaster Prevention Methods for Lifeline Systems and Task Committee (F) participation in the Third U.S. National Conference on Earthquake Engineering, August 1986, Charleston, South Carolina.

b. Workshop on Repair and Retrofit, Task Committee (C), to be held, if possible, in conjunction with the 19th UJNR Joint Meeting.

c. Fourth Workshop on Evaluation of Seismic Resistance of Existing

Buildings, Task Committee (D), to be held in Japan prior to the

19th Joint UJNR Meeting.

d. Second Workshop on Tsunami and Storm Surge Workshop, Task

Committee (I), to be scheduled in conjunction with the 19th UJNR Joint Meeting.

e. Third Bridge Workshop held by Wind and Earthquake Engineering

for Transport Systems, Task Committee (J), to be held in conjunction

with the 19th Joint UJNR Panel Meeting or in conjunction with the completion of Honshu Shikoku Bridges in October 1987

f. Workshop on Strong-Motion Earthquake Observation, Task Committee (A), to be held in the San Francisco area, August 1987.

XVII
Scheduling for workshops amd conferences shall be performed by the chairmen

of each Task Committee with concurrence of the Joint Panel chairmen. Results of each workshop shall be presented at subsequent Joint Panel

Meeting.

5. The Panel endorses the recommended change in Task Committee (E) title "Land Use Programs for Controlling Natural Hazard Effects" to reflect the Task Committee's full scope, "Natural Hazard Assessment and Mitigation through Land Use Programs".

6. The Panel recognizes the importance of continued exchange of personnel, technical information, research data and recorded data on significcuit earthquakes and strong winds, and use of available facilities in both

countries. Thus, these activities should be strengthened and expanded. To facilitate the exchange of personnel, the Panel will continue to encourage such exchanges and provide official Panel sponsorship.

7. The Panel reccannends that for future Joint Panel Meetings, priority

attention be given to the following items:

a. Task Ccxnmittee Chairmen should give attention to identifying joint reseeurch needs, priorities, development plans, and reconmendations for future research programs. Each Task

Ctaianittee Chairman is encouraged to examine his Committee's objectives and to modify the objectives, as appropriate, to meet the need for improved technology.

b. Task CCTomittee Chairmen are encouraged to increase communication

with each other for improved understanding of Panel activities. Task Conmittee Chairmen should furnish copies of appropriate Task Conmittee correspondence to the Secretariat of both sides for proper dissemination.

8. Recognizing that wind and seismic disaster mitigation plans benefit

disaster-prone countries worldwide, that many developing countries

suffer from natural disasters such as extreme winds, storm surges,

e2irthquakes and tsunamis, emd that both the U.S. and Japan have been

involved in international technical cooperative programs, the Panel will continue efforts to develop and coordinate projects which provide aid and training to developing countries and maintain the exchange of technical information.

9. The Nineteenth Joint Meeting of the UJNR Panel on Wind and Seismic Effects will be held at the Tsukuba Science City, Japan, May 1987. Specific dates, program, emd itinereury will be proposed by the Japein Panel with concurrence by the U.S. Panel.
THEME I

Wind Engineering
'-

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,i>%-;>:r: s»,-^-344%««*(.'- >4^'

^ K "4^'k. « ijN?«#y«l

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.*>'(» :' «• ..;f«tii«fRO :. -i •.••.'j »U»*«4

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|'?V.. -A’ V'HfWAi. ,;!?: ,^Tr- *M’' ;., ? “ ‘ .. 4'",%, ,V

•*' .,. 4 V »*’
THE EFFECTS OF LARGE-SCALE UPSTREAM OUSTING ON THE AEROELASTIC BEHAVIOR OF SECTION MODELS

BY

D. R. Huston\ H. R. Bosch^ and H. Tanaka^

ABSTRACT

This paper will describe the results of a recent wind tunnel study which examined the effect of large-scale upstream gusting upon the aeroelastic behavior of bridge deck section

models. The large-scale upstream gusts were created with a flapping-airfoil gust generator.

The results of the section model studies demonstrate that the method of active gust generation can produce large-scale upstream

gusts which are capable of inducing measurable

changes in the flutter stability and buffeting

response of bridge deck section models. The changes which were induced in the aeroelastic

behavior of the section models by upstream

gusting will be presented along with an interpretation of these preliminary results.

KEYWORDS: Aeroelasticity; Large-scale; Section

model; Turbulence

1. INTRODUCTION

The interaction of a flexible structure and a moving fluid, such as a large suspended-span

bridge subjected to wind loading, is an inherently complex problem. The complexities

arise because the fluid and structural motions

are coupled by the continuity of displacements

and tractions at the fluid-structure interface.

The presence of turbulence in the flow introduces an additional complication to the interaction problem. An understanding of fluid- structure interaction problems usually requires

the introduction of mildly restrictive

theoretical assumptions which are then used to formulate parsimonious semi-empirical models

with easily measured parameters.

The flexibility and exposure to the wind of the

deck structure on a long suspended-span bridge

enables an analysis of the wind loading to be conducted by considering only the action of the wind loading upon the flexible bridge deck.

If, in addition, "strip-theory" aerodynamics is valid for the bridge deck structure, then it is possible to examine the effects of wind loading

on the full bridge structure by studying the dynamics of a conceptual two-dimensional slice

of the bridge deck, i.e. a section model. Section models are extremely useful in the theoretical modeling of the aeroelastic

behavior of elongated structures. The simplicity of section models also facilitates

their use in wind tunnel simulations.

The equations of motion for a two-degree-offreedom section model take the form:

M(h + 2 ^j,u)h h ~ -L(t) [ 1 ]

I [a + 01,2 at + = M(t) [ 2 ]

where

h(t) = the vertical displacement of the section model

ci(t) = the torsional displacement of the section model

L(t) = the lift

M(t) = the aerodynamic moment

The generalized aerodynamic forces are theoretically calculated by integrating the instantaneous local surface pressures which act on the deck section as the section undergoes a virtual vertical or torsional displacement.

The fluctuating aerodynamic pressures and the associated generalized forces depend upon the upstream flow conditions, the deck section

geometry, the motion history of the deck section and local topographical considerations.

Since the theoretical calculation of the local aerodynamic pressures requires an integration

1 Dept, of Civil Engin., The Johns Hopkins

University, Baltimore, MD 21218

2 Federal Highway Administration, HNR-10, 6300 Georgetown Pk., McLean, VA 20212

^ Hitachi-Zosen Corp., 3-40, Sakurajima

1-chome, Konohana-ku, Osaka 554, Japan

3
of the partial differential equations of the unsteady fluid motion around the bluff body, it is usually necessary to invoke additional simplifying assumptions so that the fluid forces can be faithfully modeled. A primary

simplifying assumption is to model the fluctuating aerodynamic forces by conceptually

separating the aerpdynamic forces into two components, one which depends on the motion of the deck section the other being a buffeting

force . which depends on the upstream gusting,

i.e.

L(t) = L„{t) + Lb(t)

M(t) = Mm(t) + Mb(t)

where the subscripts m and b correspond to the motion-dependent and the buffeting aerodynamic

forces, respectively.

Stability of dynamical systems, such as bridge

decks subjected to wind loading, about positions of equilibrium can be assessed from the linearized motion-dependent forces acting

on the system in the neighborhood of the equilibrium position, ref. (1). The motiondependent aerodynamic forces are usually expressed in terms of the flutter derivative coefficients as follows, refs. (2) and (3):

L^(S) = pu2b [khJ (K) 4’ - KH;(K) a>

^ k2 h; (K)a + k2 h; (K) -h] [5]

M„,(S) = pu2 b2 [kA* (K) + K a* (K)o' [-g] + r2 a* (K) a + r2 a* (K) -g-J

where s and K are the reduced time and frequency, respectively.

The aeroelastic stability of a bridge deck depends upon whether the combined action of motion-dependent aerodynamic and structural

forces will cause a net increase or decrease in the total energy of the system. If the system

acts to input energy from the airstream, then

the deck motion will become excessively large

and eventually result in damage to the structure. Otherwise, the aeroelastic decksection system will dissipate the energy contained in incipient motions, thereby forming

a stable system. The calculations which assess

the aeroelastic stability of the deck sections

and the corresponding prototype bridge structures are usually iterative procedures

that search for the lowest mean velocity

corresponding to a neutrally stable system. Experiments conducted with bluff bodies indicate that aeroelastic instabilities are usually of the single-degree-of-freedom type which are associated with a negative effective

aerodynamic damping. The effective torsional

and vertical aerodynamic damping are represented in terms of the a* (K) and the flutter derivative ^ coefficients, respectively. Positive trends in either of these flutter coefficients are indicative of destabilizing tendencies. Therefore, a measurement of the a$ (k) and the h* (k) flutter derivatives will usually give a very

good indication of the tendencies towards bluff-body flutter.

If it is assumed that the motion-dependent

aerodynamic forces are indeed independent of

the upstream wind gusts, then the stability of the aeroelastic system about a position of equilibrium will be independent of the presence

of upstream turbulence and the flutter derivatives will not change under smooth and turbulent flow conditions. However, if the aeroelastic stability is affected by upstream

gusting, then the flutter derivatives will be noticeably different under different flow conditions. Flutter derivatives are usually measured with

initial conditions experiments that extract the effective aeroelastic damping from the response

that ensues from giving the section model either an initial torsional or vertical displacement.

The buffeting effect of upstream turbulence is to produce generalized aerodynamic forces that are presumably independent of the deck section

motion. The turbulent wake downstream of the bluff body, i.e. the signature turbulence, will

also produce a buffeting effect, regardless of the presence of upstream turbulence. The combination of upstream turbulence with signature turbulence will presumably produce

larger buffeting forces than those which form

in smooth flow which are solely attributed to the signature turbulence. Wind tunnel simulations of the prototype

turbulent flow conditions are usually hampered

by problems of scale. The wind gusts encountered at the bridge site are, in general, several times larger than the width of a typical bridge deck, ref. (A). The gusts which

can be simulated in a wind tunnel by passive

methods such as grids, strakes, wooden blocks.

4
etc. are usually somewhat smaller than the typical bridge deck section model, ref. (5). Two alternative solutions are available for the gust scale simulatioh problem. One alternative

is to use significantly smaller section models,

ref. (6). The use of small section models will alleviate the gust scaling problem but will exacerbate Reynolds number scaling problems

and problems associated with faithfully

reproducing the geometry of the prototype deck sections. The other alternative solution to the gust scaling problem is to use active gust generation techniques such as flapping airfoils

or air-jet injectors.

This paper will present some of the results of

a wind tunnel investigation of turbulent-wind

bridge aeroelasticity. The section models were tested under under smooth and gusting flow conditions. The gusting flow conditions were passively created, in one set of tests, by a grid and actively induced, in another set of

tests, by a flapping-airfoil gust generator.

The flapping-airfoil gust generator is capable

of producing gusts that are several times

larger than the width of a deck section.

The following sections will discuss, in succession, the experimental procedures and apparatus, the results of some of the experiments and a preliminary interpretation of the experimental results.

2. EXPERIMENTAL APPARATUS AND PROCEDURES

The innovative apparatus in this project were

the flapping airfoil gust generators. The gust generators are positioned upstream of the deck section and create two-dimensionally coherent

gusts by flapping an array of airfoils in unison. Two gust generators were used in this manner. One generator, referred to as the in- phase gust generator, flapped all of the airfoils in one direction, thereby controlling

vertical gusts. The other generator, referred

to as the out-of-phase gust generator flapped

two airfoil arrays in opposite directions, thereby tending to choke off the flow and controlling the horizontal gust action.

The use of flapping-airfoil type gust generators for wind engineering purposes was originally proposed by Scanlan, ref. (7). The development of the gust generator technology

was conducted at the Colorado State University

Fluid Dynamics and Diffusion Laboratories, ref.

(8). The result of these studies was the fabrication and installation of flappingairfoil gust generators in the Federal Highway Administration's G. S. Vincent Wind Tunnel,

McLean, VA. Details of some of the flow patterns that can be produced with the active

gust generators can be found in ref. (9). A schematic diagram of the flapping-airfoil gust generator mechanism is shown in fig. 1.

The experiments that are reported in this paper represent the first use of the flapping-airfoil

gust generators, described above, to the testing of bridge deck section models. Therefore, the experiments were designed so that the influence of the gust generators could

be easily detected and evaluated. This design entailed performing standard state-of-the-art

testihg procedures on section models in both smooth ahd gusting flow patterns.

Two types of smooth flow patterns were employed

in this study. One type was formed when the gust generators were removed from the wihd

tunnel. The resulting undisturbed air stream

was quite smooth with a turbulence intensity of less than 1%. This approximately laminar flow will be referred to as the "Laminar" flow. The other type of smooth flow was produced, while

the gust generators were installed in the wihd

tunnel. This flow was created by passing the air through the geherators with the airfoils

held in a neutral stationary position. The resulting flow was reasonably smooth flow with

an intensity of approximately 3%. The flow produced by holdihg the gust generator airfoils

in a neutral position, will be referred to as the "Neutral" flow. Large-scale gusting flow patterns were induced

by flapping the airfoils on the in-phase gust generator with a large-amplitude, 2.0 Hz. cutoff, low-pass filtered white noise random driving signal. This driving signal was selected so that the gust generators would

produce produce one of the more violently

turbulent flow patterns that the gust geherators were capable of syhthesizing. The gusting flow, which formed downstream of the generators and upstream of the section models, with this driving signal, was not intended to simulate any specific prototype field condition, but rather it was simply a large- scale high intensity flow pattern. Since the flow was produced by driving the in-phase gust generator with a 2.0 Hz. low-pass filtered white noise random signal, hereinafter it will

be referred to as the "Low-Pass" flow. The intehsity of the turbulence contained in the "Low-Pass" flow was measured at 20%. The lehgth of the resultihg gusts was quite large and could be extended, in theory, to infinity.

An additional type of turbulent flow was created by placing a 4x4 wood lath grid across

the mouth of the wind tunnel. The wooden grid induced turbulence intensities that ranged from

10% to 20%. The flow patterh produced by the wooden grid will be referred to as the "Grid Turbulent" flow. The scale of the "Grid" turbulence and the spanwise coherence was somewhat smaller than that of the "Low-Pass"

flow.

5
Three different section models were tested under these four flow regimes. The^ deck section models were geometrically faithful slices of the Deer Isle-Sedgewick, Maine, bridge, a proposed fairing-modified version of the Deer Isle bridge and the Golden Gate

Bridge, San Francisco, Calif., figs. 2.a-c. The unmodified Deer Isle bridge is of the plate

girder or H-type. The proposed fairing- modified version of the Deer Isle bridge forms

a rather streamlined deck section. The Golden Gate deck section is of the open-truss type. The standard section model tests consisted op

1.) measuring the motion-dependent aerodynamic

forces, i.e. the flutter derivatives, by initial condition experiments, 2.) recording

the stationary response of the section model to stationary wind conditions. Both sets of experiments were performed with the deck section models mounted on flexible springs. This elastic support structure was constructed

to permit motion in the vertical direction and

in a torsional direction about a spanwiserunning axis. The support structure was designed to restrain or enable the vertical and

the torsional degrees of freedom separately, if desired. Therefore, the section model could be configured as either a single-degree-of-freedom

system or as a two-degree-of-freedom system. All of the flutter derivatives were measured

with the section model configured to act as a single degree of freedom system. The flutter derivatives that could be measured with an S.D.O.F. setup were measured were the A|(K), A*(K), Hf(K) and the H*(K) flutter derivatives,

i.e. the auto-excited flutter derivatives. The stationary time histories were recorded with

the deck sections mounted in three different

configurations, one which permitted single

degree of freedom vertical motion, one which permitted single degree of freedom torsional motion and one which permitted two degree

_ of freedom combined vertical and torsional motion.

3. EXPERIMENTAL RESULTS

3.1 Flow Measurements

The ability of the gust generators to produce controllable gusts is demonstrated by figs. 3-

5. Figs. 3. a and 3.b are strip chart recordings of the flow that resulted from holding the airfoils in a stationary neutral

position, i.e. the "Neutral" flow, at a mean velocity of 7 ft. /sec, with an intensity of about 3%. An observation of these traces indicates that the "average" gust size is of the order of 1 ft. or less. Figs. A. a, A.b and A.c are traces of the actuator motion and the gust velocities that result from flapping the in-phase airfoils with a 2.0 Hz. low-pass

filtered, large-amplitude, white noise signal. The intensity of the resulting turbulence was

20%. An observation of figs. A.b and A.c

indicates that the "average" gust size which

results from using these flow patterns is at least 7 ft, which is several times larger than

the 1ft. -2ft. range^of typical section model

widths. Spectra of the actuator motion, vertical gust velocity, and the horizontal gust velocities which correspond to the "Low-Pass"

flow are shown in fig. 5. 3.2 Flutter Derivative Measurements

The A2;(K), A*(K), H*(K), and H*(K) flutter derivatives were extracted from the three section models under smooth and gusting flow patterns. fairing-modified Deer Isle deck section and the Golden Gate deck section were

all measured under smooth and gusting flow conditions. Both versions of the Deer Isle deck section were tested under the "Laminar," "Neutral," "Grid Turbulent," and "Low-Pass"

flows. The Golden Gate deck section was tested only under the "Neutral" and the "LowPass" flows. The flutter derivatives which were measured by initial condition experiments are shown in figs. 6-8. Figs. 6,a-6.d contain the flutter derivatives that were extracted from the unmodified Deer Isle deck section. Figs. 7.a- 7.d are the flutter derivatives which were extracted from the fairing-modified version of the Deer Isle deck section. Figs. 8.a-8.d are

the flutter derivatives which were extracted

from the Golden Gate deck section.

All of the flutter derivatives that are shown

in figs. 6-8 were calculated from estimates of the effective aeroelastic damping and natural

frequency. These estimates were performed

with a digital version of the logarithmic

decrement procedure that used all of the local maxima and minima in the free vibration

response of the deck section to calculate the desired system parameters, ref. (10). Since the logarithmic decrement procedure is strictly applicable to the case where a linear

system is oscillating as the result of nonzero

initial conditions without any applied external

forces, such as buffeting forces, it was decided to verify some of the measured flutter derivatives with an independent procedure. One alternative method of estimating damping ratios

and natural frequencies of elastically

supported section models, is to measure the width of the peak that occurs in the spectrum

of the displacement that results from buffeting

the section with a broadbanded stationary random excitation. A modified version of this procedure which used a least-squares spectral

curve fitting procedure, ref. (10), coupled

with the use of a Maximum Entropy spectral

estimator, ref. (11), was used to remeasure the A*(K) flutter derivative under smooth and gusting flow conditions. The results of this independent flutter derivative extraction

6
procedure to the stationary response of the Golden Gate deck section are shown in fig. 9. Since it is impractical to measure the reference zero-velocity viscous damping ratio

by the spectral width procedure, the vertical positioning of the horizontal axis in fig. 9 is uncertain. If the mean of viscous damping

ratios where were measured with transient

initial condition experiments was employed,

then the horizontal-axis takes the position of

the solid line in fig. 9. However, since the

A*(K) flutter derivative should approach zero

as the reduced velocity approaches zero the horizontal axis in fig. 9 should be shifted to reflect this physical constraint. The dashed horizontal axis which appears in fig. 9 reflects this ad hoc axis shift, which is based

on physical reasoning. 3.3 Stationary Buffeting Response

The motions which resulted from allowing the elastically-supported section models to oscillate freely in the presence of buffeting

forces due to upstream and signature turbulence

were digitized and recorded on magnetic tape. Figs. 10-12 contain plots of the dimensionless

root mean square displacement versus the reduced velocity for four different motion

conditions. The torsional displacement is expressed in terms of radians, the vertical

motion is rendered nondimensional by dividing

the motion by the deck width. The wind velocity is reduced by the free vibration

natural frequency of the particular degree of freedom under consideration and by the deck

width.

The use of three different end support conditions required the observation of four different types of motion. The four displacement-types were: 1.) vertical singledegree-of-freedom motion, 2.) torsional singledegree-of-freedom motion, 3.) the vertical component of a two-degree-of-freedom motion,

and 4.) the torsional component of two-degree- of-freedom motion. Figs. 10. a. -c. contain the RMS vs reduced velocity plots for the unmodified Deer Isle deck section under four

flow conditions. Fig. 11. a. -d. contain the rms displacement vs. reduced velocity plots for the fairing-modified Deer Isle deck section. Figs.

12. a. -12. d. contain the RMS displacement vs. reduced velocity plots for the Golden Gate deck

section.

4. INTERPRETATIONS AND CONCLUSIONS

The measurements of the flow conditions

downstream of the gust generators indicated

that the flapping airfoils are indeed capable

of producing controllable large-scale gusts.

If the airfoils are held in a neutral stationary position the resulting turbulence

flow field is probably smooth enough to be used

as a nongusting reference flow condition. The ability of the gust generators to simulate prototype field conditions has not been entirely verified, although these studies indicate that it should at least be possible

for those prototype conditions where the turbulence intensity exceeds 10%. The flutter derivative measurements indicated

that the presence of upstream turbulence will produce noticeable changes in the flutter derivatives. The AJ(K) flutter derivative for the unmodified Deer Isle deck section displayed

a moderate change under smooth and gusting flow conditions. The A*(K) derivative for the "Laminar" flow was unmeasurable in the reduced

velocity range of 2-4 because of strong vortexshedding action. The results indicate that the presence of the upstream turbulence in the "Low-Pass" flow have indicated causes a slight

positive trend in the A*(K) derivative, which

indicates a destabilizing tendency due to the upstream gusting. The fairing-modification to the Deer Isle deck section produced radically

different flutter derivatives from those of the unmodified Deer Isle deck section. Nevertheless, the presence of upstream gusting

in both the "Grid Turbulent" and the "Low-Pass"

flow causes an increased positive

(destabilizing) trend in the A*(K) derivative. Likewise the Aj(K) derivative for the Golden Gate Deck section also showed this same destabilizing trend in the A*(K) derivative for the smooth and gusting flows. Since the tendency of upstream gusting to produce an increased aeroelastic torsional instability is somewhat contrary to the conventional wisdom that turbulence is a stabilizing factor in bluff-body aerodynamics,

the AJ(K) flutter derivative for the GoldenGate deck section was also measured by the alternative least-squares spectral width

procedure. The remeasured A*(K) derivative

shows the same destabilizing tendencies due to upstream gusting as that of the transient

response measurements. Therefore, in addition

to verifying the result that upstream gusting

can indeed exert a destabilizing influence upon

the torsional dynamic motion of bridge deck section models, the two methods of system identification were verified.

The RMS displacement vs. reduced velocity curves exhibited a behavior that was consistent

with the results of the flutter derivative

experiments. The unmodified Deer Isle deck section showed an increased RMS response due to the upstream gusting except for certain

isolated peaks which may be attributed to the vortex-shedding action in the smoother

"Laminar" and "Neutral" flows. The changes in the flutter stability of the flutter stability

of the unmodified Deer Isle deck section were

not extremely dramatic. The fairing-modified

Deer Isle deck section showed a considerable

7
increase in the vertical motion due to the presence of upstream turbulence. Since the addition of fairings produced a deck section

that was quite stable, no changes were observable in the aeroelastic stability of the fairing-modified Deer Isle deck section. The Golden Gate section also exhibited an increased vertical response to upstream gusting. The presence of turbulence in the "Low-Pass" flew caused to section to become unstable at a lower velocity than that which was encountered in the smoother "Neutral" flow. This destabilizing

trend was verified by the flutter derivative

measurements. The instability which appeared

in the Golden Gate motion was a torsional flutter-type instability, in spite of rather larger vertical buffeting motions. Some of the overall conclusions that can be drawn from this study are that the use of flapping-airfoil type gust generators is a viable method of producing controllable largescale gusting flow patterns. The future research capabilities of this apparatus are virtually unlimited. The measured chahges in the aeroelastic behavior of the section models,

at present forms a rather small data base. However, the preliminary conclusions that can

be drawn from these results are that large- scale upstream gusting may, in some circumstances, be destabilizing and the gusting

will possibly suppress vortex-shedding action.

ACKNQVjLEDGMENTS This research was conducted at the U. S. Department of Transportation, Federal Highway

Administration, Turner-Fairbank Highway Research Center's wind tunnel facilities in McLean, VA. During the course of this study

the first author received financial assistance

from the National Highway Institute and the third author received financial support from

the International Road Federation. REFERENCES

(1) Meirovitch, L., Methods of Analytical

Dynamics , McGraw-Hill, 1970.

(2) Scanlan, R. H. and Tomko, J. T., " Airfoil

and Bridge Deck Flutter Derivatives" ASCE,

Jnl. Eng. Mech. Div., pp. 1717-1737, Dec. 1971.

(3) Beliveau, J. G. Suspension Bridge

Aeroelasticity Nonlinear Least Squares Techniques for System Identification , doctoral dissertation, Dept. o? Civil Engineering, Princeton University, Sept.

1973.

(A) Counihan, J. "Adiabatic Atmospheric

Boundary Layers: A Review and Analysis of Data from the Period 1880-1972" Atmos. Environ., 9(1975), 871-905.

(5) Jancauskas, E. 0. and Melbourne, W. H., "The Aerodynamic Admittance of Two- Dimensional Rectangular Section Cylinders

in Turbulent Flow," Proc. Fifth U.S. National Conference on Wind Engineering,

Lubbock, Texas, 1985.

(6) Lin, W. H., Forced and Self-Excited Responses of a Bluff Structure in a Turbulent Wind , Ph.D. dissertation, Princeton University, 1977.

(7) Bienkiewicz, B., Cermak, U. E., Peterka,

J., and Scanlan, R. H., "Active Modeling of Large-Scale turbulence," J. Wind Eng. Ind. Aerodyn,. 13(1983), A65-476.

(8) Cermak, J. C., Bienkiewicz, B. and Peterka,

J. , Active Modeling of Turbulence for Wind

T unnel Studies or Bridge Models, Final Report FHWA/RD-82/148.

(9) Huston, D. R., and Bosch, H., "Active Modeling of Turbulence for Section Model

Studies," Presented at the 16th U.S. -Japan

Joint Panel on Wind and Seismic Effects, National Bureau of Standards, Gaithersburg,

MD, May 1984.

(10) Huston, D. R., The Effects of Upstream

Gusting Upon the Aeroelastic Behavior of

Long Suspended-Span Bridges , doctoral dissertation, Princeton University, Department of Civil Engineering, May 1986.

(11) Robinson, E. A., Multichannel Time Series Analysis With Digitil Computer Programs , Holden-Oay, San Francisco, 2nd Ed., 1978.

A Horizontal Gust Control B • Vorticol Gu«t Control C Linkogo 0 Hydraulic Actuators E * Bridge Modal

Fig. 1. Line drawing of the flapping-airfoil

gust generators positioned downstream of the wind tunnel exhaust nozzle and upstream of the section model.

8
(7

Fig. 2. a. Unmodified Deer Isle-Sedgewick,

Maine, bridge deck section.

Fig. 2.b. Fairing-modified version of the Deer

Isle deck section.

9
a. Horizontal gusting.

Fig. 3. Typical simultaneous gust velocity time histories for the "Neutral" flow at a nominal mean velocity of 7 ft. /sec.

Fig. 4. Typical simultaneous gust velocity time histories for the "Low-Pass" flow at a nominal mean velocity of 7 ft. /sec.
SEC.

c. Vertical gusting.

Fig. 4. (cent.) Typical simultaneous gust velocity time histories for the "Low-Pass" flow at a nominal mean velocity of 7 ft. /sec.

Actuator Displacement Spectrum

Fig. 5 . Actuator and gust velocity spectra corresponding to the "Low-Pass" flow.

11
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17
ON THE EVALUATION OF WIND-INDUCTED VIBRATIONS OF LONG-SPAN

BRIDGES USING TAUT-STRIP MODELS

Nobuyuki Narita*, Hirohiko Tada*, Koichi Yokoyama*, Hiroshi Sato*, Kazuo Kanzaki*, and Makoto Fukuda*

ABSTRACT

The taut- strip model test is a new wind tunnel

technique. A bridge deck model is attached to two parallel taut piano wires, and its responses to wind are measured. The advantages of

the taut- strip model test over the section

model test are that the model can be tested in turbulent boundary layer flow, and that the model is three dimensional. Therefore, the properties of wind and bridge vibrations for taut-strip model tests are more similar to prototype than those for section model tests.

Also, the construction cost of the taut-strip

model is less expensive than that of full bridge model. This paper describes the design method of the taut-strip model, the simulation method of turbulent boundary layer flow, and the test results on the wind- induced vibrations of the taut-strip models. The flutter speed of the flat plate model was compared with the one calculated from Selberg's formula, and the vibrations of the box girder model were

compared with those observed at the real cable stayed bridge. It was confirmed that the taut- strip model test is one of the effective

techniques for evaluation of the wind- induced vibrations of long- span bridges with box girder

deck.

1. INTRODUCTION

The wind-resistance of long-span bridges in Japan have been investigated and proved through

the section model tests in smooth flow. Natural wind, however, is turbulent flow, whose wind speed varies with time and space, and the vibrational displacement of bridge deck varies

along the bridge axis. Therefore, the wind- induced vibrations of long- span bridges can not

be predicted directly from the section model

tests in smooth flow. On the other hand, the wind- induced vibrations can be obtained

directly through the full bridge model tests in turbulent boundary layer flow if the structural

properties and flow properties are kept similar

to those of prototype. The construction of

full bridge model, however, is more expensive

and difficult than that of section model.

Recently, a new technique called taut-strip

model test has been developed (1,2). In taut- strip model tests, the bridge deck model is attached to two parallel taut piano wires, and

its responses to wind are measured. The advantages of the taut-strip model test over

the section model test are that the model can

be tested in turbulent boundary layer flow, and

that the model is three dimensional. Therefore, the properties of wind and bridge vibrations for taut-strip model tests can be kept

more similar to prototype than those for sec- tion model tests. Another advantage is the construction cost of the taut-strip model is less expensive than that of full bridge model. This paper describes the design method of the taut-strip model, the simulation method of turbulent boundary layer flow, and the test results on the wind- induced vibrations of the taut-strip models. The flutter speed of the flat plate model was compared with the one calculated from Selberg's formula, and the vibrations of the box girder model were compared with those observed at the real cable stayed bridge.

2. DESIGN METHOD OF TAUT-STRIP MODEL

2 . 1 Similarity Requirements

In the wind tunnel test of long- span bridges,

the following nondimens ional parameters should

be kept the same between prototype and model.

cr-m/CpB^), 7»I/(pB''), U/(fhB), U/(f^B), f^/fh

where, m:

P'

B:

I:U: fh =

«h:

5 ^:

mass per unit length of the bridge, air density

bridge deck width

mass moment of inertia per unit length of the bridge mean wind speed

natural frequency of vertical bending

mode natural frequency of torsional mode logarithmic decrement of vertical bending vibration

logarithmic decrement of torsional vibration

*Public Works Research Institute, Ministry of To simulate the natural wind, the following

Construction, Tsukuba, Japan. parameters should be kept the same.

18
UZ2/UZ1. lu. Iw. ki/B. W/B fh - (21)-i(2P/m)^

where

U^x ; mean wind speed at the height of Z1 U22 • mean wind speed at the height of Z2 turbulence intensity of

u( longitudinal component of fluctuating wind speed)

I„; turbulence intensity of w(vertical

component of fluctuating wind

speed)

1^: integral scale of turbulence of u

integral scale of turbulence of w

The structural dimensions of the taut-strip

model can be determined by the following

process

a. The scale ratio should be decided

considering the width of the test section

of the wind tunnel, span length of the bridge, the integral scale of turbulence of

the natural wind at the construction site, and that of the boundary layer generated in the wind tunnel.

fg - (b/(41)) (2P/I)^(l+(2GJ)/(PbM)^

fff/fh - (b/2) (m/I)^(l+(2GJ)/(Pb2))^

where, P: tension per one piano wire

1: length of the taut-strip model

b: separation of the two piano wires

GJ: torsional stiffness of the bridge deck model

f^ can be adjusted by P, and fg/f^ can be adjusted by GJ and b. When the each bridge deck segment is attached

by point, the formulae can be rewritten as follows

f^, - (21)-i(2P/m)^

fff - (b/(41)) (2P/I)^

f^/fh - (b/2) (m/I)^- (b/(2B)) ( 0/ 7 )^

b. The required values of m and I should be calculated using the scale ratio.

c. The natural frequencies fjj and fg should be decided so that the wind- induced vibrations

for the required wind speed range may be investigated in the wind tunnel. Attension

should be paid to the sag when small values

of fjj and fff are applied.

To keep the piano wires inside the deck model,

b should be less than B. Therefore fg/f^ is less than 0. 5 ( 0/7 )^. If the required fj/fh is larger than this value, the taut-strip model should be constructed so the torsional stiff- ness of the bridge deck model may contribute to the whole taut-strip model.

3 . SIMULATION OF TURBULENT BOUNDARY LAYER FLOW

2.2 Natural Frequency 3 . 1 Simulation Method

Since the required values for m and I are given

from similarity requirements, the natural

frequencies should be adjusted by changing the stiffness of the taut-strip model. The stiffness of the model is given by the taut piano wires and the bridge deck model. When

the bridge deck model is separated into several

segments and attached to the taut piano wires,

the bending stiffness of the bridge deck model

does not contribute to that of the whole taut- strip model. Only the stiffness due to the piano wires should be considered. For the torsional stiffness, the contribution of the bridge deck model depends on the length where

the road deck model is attached to the piano

wires. If each deck segment is attached by

point, the torsional stiffness of the deck

model can be neglected. In the case that the length of the attached part is not negligibly

small, the effect of the torsional stiffness of

the bridge deck should be considered.

The natural frequencies f^j and fg of the taut- strip model can be calculated by the following

formulae when the bridge deck model segments

are fully attached to the taut piano wires.

For the simulation of turbulent flow, the turbulence properties of natural wind at the bridge construction site should be investigated. In the case that field observations can

not be made, the properties may be estimated

from the references (3,4,5). Among many non-dimensional parameters for similarity requirement, the power law exponent

Q was chosen as a fundamental parameter

because it reflects the effect of the terrain

roughness in a simple manner. Typical values

of a for several terrain conditions are as follows (3,4). Rough Sea a-0.12 Open Grassland a-0 . 16 Suburban Areas a-0. 28 City Centres a-0. 40

In the wind tunnel, turbulent boundary layer

flow can be generated by floor roughness and

spires. The relation between the shape of spires and a is shown in the reference (6)

Appropriate floor roughness should be chosen to make a a required value.

19
3 . 2 Experiment Since

The low speed wind tunnel of the Public Works Research Institute, with test section of 2.5m

wide, 4m high and 10m long, was used for the experiment. The four corners of the test section have been cut by 0.4m, and the width of

the tunnel floor is 1.7m. To use the full width of the test section, the surface floor

was installed 0.4m above the wind tunnel floor

as is shown Fig. 1. At the upstream end of the surface floor, the fairing plate was attached

to prevent the separation of the flow. In the case of the generation of smooth flow, a shorter surface floor was used as is shown in

the table of Fig. 1. The aim of the experiment was the generation of turbulent boundary layer flows whose power law exponents correspond to the natural wind over

rough sea (turbulent flow I), open grassland

(turbulent flow II) and suburban area (turbulent flow III). The arrangement of the spires

and the floor roughness for each turbulent flow

is shown in Fig. 1. The measurement was made

by hot-wire anemometer.

3.3 Results

3.3.1 Wind Profile

The vertical profiles of the mean wind speed of

the turbulent boundary layer flow are shovm in

Fig. 2. The power law distributions are also plotted in the figure. Although the exponent

of the turbulent flow III is slightly higher

than the typical value for suburban areas, the wind profiles of the turbulent flow generated

in the wind tunnel are fairly similar to those

of the natural wind.

3.3.2 Properties of the Turbulence

The power spectral density functions (PSD) S of

the fluctuating wind speed were calculated and reduced by frequency f and mean wind speed at

the height of Z, U^. The reduced PSD, are plotted against reduced frequency fB/U^

(where B is bridge deck width) in Fig. 3. The values of Z and B for the experiment are 0.14m

and 0.0925m, respectively, which correspond to the height and the deck width of the 1/200 bridge deck model, whose wind- induced vibrations are described in Chapter 4. The PSD's of natural wind for typical surface conditions are

also plotted in Fig. 3, using the Engineering

Science Data Unit (5). The surface roughness

parameters zq were assumed as follows (5). Large Expanses of Water

Fairly Level Grass Plains Outskirts of Towns

Centres of Large Towns

zq-0 . 001m

zq-0 . 01m

zq-0 . 4m ZQ-lm

fS/Uz^-iZ (fs/a2) fg/Uz-B/L (Lf/Uz) where a is standard deviation of fluctuating - wind speed, the extent of similarity about turbulence intensity I and scale of turbulence

L can be judged by comparing the plotted PSD of

the experiment with that of the prototype. From the figure, it is shown that the turbulence scales are almost similar between model

and prototype when the scale ratio of the model

is about 1/200. As for the turbulence intensities, the turbulent flow generated in the wind

tunnel may be regarded as the model of natural wind over the following surface conditions. Turbulent Flow I Large Expanses of Water Turbulent Flow II Fairly Level Grass Plains Turbulent Flow III Outskirts of Towns or Centres of Large Towns

4. WIND- INDUCED VIBRATIONS OF TAUT- STRIP MODEL

4. 1 Cases and Conditions of Experiment

To confirm the applicability of the taut-strip

model test to the resistant design of long-span

bridges, flat plate models and box girder

models were constructed. The flutter speed of

the flat plate model can be compared with the theoretical value. The 1/200 model of the cable stayed bridge, where wind and responses

of the bridge deck have been observed (7) , was constructed so the responses of the model could

be compared with the ones observed at the bridge site. The taut-strip models are shown in Fig. 4. For

the flat plate, two models with different

frequency ratio f^/fh were constructed. For

the box girder, the model with the flap plates

was constructed as was the model with usual

handrails. The flap plate is one of the aerodynamic devices for reduction of the vortex- inducted vibrations, and it was applied

to this cable stayed bridge.

The structural properties of the taut-strip

models and those of the prototype cable stayed

bridge are shown in Table 1. Except for the logarithmic decreasement of the torsional

vibration, the structural properties of the box

girder model is fairly similar to those of the prototype

.

The experimental cases are shown in Table 1. Wind- induced vibrations were measured in the smooth flow and in the turbulent flow II. The turbulent flow II was chosen because its turbulence properties are similar to those of

the wind at the cable stayed bridge site.

In this case, the values of the prototype cable stayed bridge were used for B and Z.

20
In the experiment, the displacement at the center of the model was measured by the optical

displacement meter. The wind direction was perpendicular to the model axis, and the angle

of inclination was 0. 4 . 2 Results

4.2.1 Flat Plate Model

The relations between the vibrational

displacement and the mean wind speed are shown

in Fig. 5. The maximum and the root mean

square of the displacement were calculated

using 1 minute data, and reduced by the deck

width B. The main findings are as follows:

a. Coupled flutter was observed for the model

of small frequency ratio tested in the smooth flow (Case 1).

b. The reduced flutter speed (U^/CfgB)) was about 8, which agreed fairly well with the one (U2/( f ^B)“7 . 8) calculated from

Selberg's formula (8).

c. Buffeting of vertical bending mode and torsional mode was obseirved for the model

of small frequency ratio tested in the turbulent flow (Case 2).

d. For the model of large frequency ratio, coupled flutter or remarkable buffeting of torsional mode was not observed (Case 3,4), because of the reduced wind speed range tested was low. 4.2.2 Box Girder Model

The experimental results of the box girder

models are shown in Fig. 6 in the same manner

as those of the flat plate models. The main

findings are as follows.

a. Vortex- induced vibrations of vertical bending mode were observed in the reduced wind speed (U2/(fhB)) from 2.0 to 2.5 for

the usual handrail model tested in the smooth flow (Case 5)

b. The amplitudes of the vibrations were reduced remarkably by attaching the flap plate (Case 7)

c. In the turbulent flow, no vortex- induced vibration was observed, but the buffeting

of vertical being mode was observed for the usual handrail model and for the flap model

(Case 6,8). Since the experimental condition of Case 8 was regarded as the most similar to the prototype

cable stayed bridge, its responses were compared with those observed at the bridge.

The results are shown in Fig. 7. At the bridge, wind and response data have been

collected up to the mean wind speed of 30 m/s, and responses are regarded as the buffeting of vertical bending mode. As far as this wind

speed range is concerned, the responses estimated from the taut-strip model test agree fairly well with those observed at the prototype cable stayed bridge.

5. CONCLUSIONS

The taut-strip models for flat plate and box

girder deck of cable stayed bridge were

constructed, turbulent flow was simulated, and

their responses to the smooth flow and the turbulent flow were measured. The major

findings are as follows:

a. The structural properties of the taut-strip

model can be kept almost similar to those

of the prototype bridge deck, except for

the logarithmic diminishing of torsional

mode of the large frequency ratio model.

b. The turbulent flow generated in the wind

tunnel were found to be applicable to the taut-strip model whose scale ratio is about

1 /200 . c. The flutter speed for the flat plate model

in the smooth flow agreed fairly well with

the one calculated from Selberg's formula.

d. The wind- induced vibration of vertical bending mode of the box girder model agreed

fairly well with that observed at the prototype cable stayed bridge.

e. From the findings above, the taut-strip

model test is on effective technique for estimating wind- induced vibrations of long- span bridges with box girder deck.

6. REFERENCES

(1) Davenport, A. G. , N. Isymov, and T. Miyata, The Experimental Determination of

the Responses of Suspension Bridges to Turbulent Flow, Proc . 3rd Int. Conf. on Wind Effects on Bldgs, and Strcts., Tokyo,

1971.

(2) Davenport, A. G. , The Use of Taut Strip Models in the Prediction of the Responses

of Long Span Bridges to Turbulent Wind,

Proc. Symp . on Flow Induced Strct. Vibrations, Karlsruhe, 1972.

(3) Davenport, A. G. , Wind Structure and Wind

Climate, Safety of Structures under Dynamic Loading, Trondheim, TAPIR, 1978.

(4) Davenport, A. G., The Interaction of Wind

and Structures, Engineering Meteorology, Amsterdam, Elsevier, 1982.

21
( 5 ) ESDU, Engineering Science Data, Wind

Engineering, Volume 1, Wind and Turbulence

Data, Engineering Science Data Unit Ltd.

,

1982.

(6) Irwin, I. P. A. H. , The Design of Spires

for Wind Simulation, J. of Wind Engineering and Industrial Aerodynamics, 7, 1981.

(7) Okubo, T.. N. Narita, K. Yokoyama and H. Sato; Field Observations of Aerodynamic

Behavior of Long- Span Bridges, Proc. Fifth

Int. Conf. on Wind Engineering, Colorado,

1979.

(8) Selberg, A. , Oscillation and Aerodynamic

Stability of Suspension Bridges, ACTA

Polytechnica Scandinavica, CII3, 1961.

22
1]JJ]JJJ]]J]]]]1]]]

23
I

[

I

[

T

24
J

]

J

J

J

J

J

J

]

]

J

J

J

]

J

1-JJ

J

J

Turbulent Flow I Turbulent Flow n Turbulent Flow ffl Fig. 2. Vertical Profiles

Longitudinal Cosponent

of Fluctuating Wind Speed Vertical Coiponent

of Fluctuating Wind Speed

Fig. 3. Power Spectral Density Functions

25
slit

was

covered

by

tape f

26
Wind-Induced

Vibrations

(CASE

1)

52

.

Wind-Induced

Vibrations

(CASE

2)
28
29

Fig.

6.2.

Wind-Induced

Vibrations

(CASE

6)
30
Coapared

with

the

Field

Observation

Result
THE RESPONSE OF TAUT STRIP MODELS: COMPARISONS OF THEORETICAL AND EXPERIMENTAL RESULTS

BY

R.H. SCANLAnI and R.D. MARSHALL^ ABSTRACT

This paper tests, by representative calculations, the applicability of certain existing

theories of flutter, vortex-shedding, and galloping to the corroborationof results found experimentally for the response of four

taut strip models in the wind tunnel. The

report shows that the theories yield plausible predictions but are very sensitive to values of input parameters. Therefore it is imperative to acquire accurate values of such parameters for any given system to which the theories are to be applied.

ence 1. Since the reported study is exploratory and not intended to be based upon further experimentation, recourse is made to existing flutter derivatives and to other information that appears in the literature

[3,4,5,6,8,91, Using these derivatives and related data, a plausible case is made for the ability of theory to corroborate the taut strip results observed experimentally.

2, ANALYTIC BACKGROUND FOR FLUTTER INSTABILITY KEYWORDS: AERODYNAMICS; FLUTTER; VORTEX

EXCITATION; GALLOPING; TAUT STRIP MODEL

1. INTRODUCTION

Experimental findings on the response of a series of 4 taut strip models of rectangular

section that were excited by wind in a wind

tunnel are summarized in reference 1. The extent to which existing aerodynamic response

theory for bluff bodies can be used to predict the observed responses has been reported in reference 2. This paper presents

certain highlights of the work of reference 2 and the main experimental findings. The ratios of width B to depth D of the taut strip models were respectively 1,2,5 and 10. Essentially what was done in the experimental

work was to excite the 4 different taut strip models each through a range of velocities

until instability of its response became

apparent, the environment consisting of either smooth flow or successively 3 dif- ferent levels of turbulent flows, realized at different heights in the boundary layer. Figures 1 through 4 portray the main experi- mental observations, which consisted of either rms vertical or torsional displacements at midspan of the models.

As is generally accepted, theories of aeroelastic response rely, except in extremely simple cases -- such as that of the thin airfoil and a few other streamlined bodies —

upon measured, rather than purely theoretical

flutter derivatives [3]. Fortunately, some limited experimental torsional damping deri- vatives exist for three of the four rectangular section models examined in referAccording to reference 4, most bluff-body

flutter is associated with single-degree-of-freedom instability in torsion. This means that, even in those cases where vertical and twist motions couple (as they do in so-called "classical" flutter), with bluff bodies it is usually the unstable

torsion mode that "drives" the flutter

motion. Under these conditions a close approximation to the situation can be real- ized via a mathematical model permitting the torsion degree of freedom only. This is written as follows (4);

I[a + = 1/2 pU^[2B^][KA2* Ba + where

I = mass moment of inertia per unit span

of bluff body (taut strip model)

a = twist degree of freedom of model

i; = torsional mechanical damping ratio of model

u = natural circular frequency of model

U = mean cross-wind velocity

B = godel width (along-wind direction) K = — = "reduced frequency"

p = air density

0) = 2im (=ijJ ) = flutter circular

frequency A* = A„*(K) = flutter damping

coefficient, function of reduced frequency

A-* = A *(K) = flutter stiffness

coefficient, function of reduced frequency

In particular, flutter occurs when the net system damping becomes zero, i.e., when

a terms on the left and right are equal:

The Johns Hopkins University

2 Baltimore, Maryland 21218 National Bureau of Standards Gaithersburg, Maryland 20899

= 1/2 oU^[2B^][KA2* ^

]

( 2 )

32
since experimentally it is found that the flutter frequency departs very little from

the natural torsional frequency (cj=uj ), then

Bu) " IT (3) Using (3) in (2) results in the net condition that holds at flutter:

(4) Thus, for a particular set of mechanical

parameters, the critical flutter coefficient

is determined. Since it in turn is

an experimental function of K , (usually given as curves of A * vs. the parameter

U/nB, where n = tu727r) the value of K for flutter can be found. This then reveals the critical value of velocity, flutter condition.

2 . 1 Application of Theory to Corroborate

Critical Velocities of Experimental Cases Involving Torsional Flutter Instability Reference 5 provides results on the torsional

flutter instability of several rectangular-section prisms, including those with B/D = 1,2, and 5. As is shown in reference 2, the data given in reference 5 for the torsional damping coefficient C may be translated into the notation used here by the formulas below:

A

*

= 2(§) r

(5)

^

- V/(B/D)

( 6 )

2n ,B« (7) Note that U as employed i^ reference 1 bears

the following relation to V of reference 5:

h vs

.

as given by reference

5, are “presented in Figs. 5 to 7. Mechanical data for the four strip models are listed in Tables 1 and 2. In these tables the following symbols are used:

B = model width (streamwise dimension)

D = model depth (vertical)

L = total model span

= total model mass

Ij. = total model mass moment of inertia

about its rotation axis

p = air density

(; = mechanical damping ratio in vertical motion

^ = mechanical damping ratio in rotation

Plots of C^Q^ vs. V presented

= vertical frequency (Hz.) n^ = torsional frequency (Hz.)

It should be noted that the mechanical damping ratios and ;; given in Table 2 are derived from” logarithmic decrements. The average values of damping ratio (^^and ? ^

)

listed in reference 1 were not used since, used as mechanical damping ratios, they did

not, in general, lead to as good a m.atch between theory and experiment as those given

in Table 2. Characteristics of the turbulent

boundary layer established in the wind tunnel

are listed in Table 3. CASE B/D = 1 The curves of vs. V for smooth and tur- bulent flow are given in Fig. 5. Clearly,

only those regions where C is greater than

zero are of interest in predicting instability. Hence attention will be focused here on such regions.

First, the criterion for oscillatory

instability is recalled. This is given by

eq. (4). Applying this here with data from Tables 1 and 2 yields

(V)cr

_Inspection of the curve (Fig. 5) of vs. V for smooth flow reveals an interruption

(probably due to erratic model behavi^or asso- ciated with instability) be_tween V=7 . 4 and

V=9.8. If, in the region of V=8, the value

of C had attained the very plausible value

of 0.45, the corresponding A^*value would

have been, according to eq. (5): A

2 *= 0.3648 which is very near the predicted ins- tability criterion. The corresponding value

of U would have been 8.64, whereas Fig.

l(b)’^ shows a strong^ measured instability somewhat earlier, at U =6.6 However, as it stands. Fig. 5 (Ref^^ 5) does not actually

provide the critical data needed. Note jthat calculation of A * in the region of V=3

,

yielding A

2 *= 0.0684, does not reveal any torsional instability, which is consistent

with Fig. 1(b) in this region.

For turbulent flow , the curve (Fig. 5) of Cv,aT reveals po_tential instability

regions 3.4iVS5.3 and 7.41.Vi20. Calculation

of A

2 *by eq. (5) in the first of these regiom reveals that the instability criterion

for A * is not met, consistent with the results of Fig. Kb). However, for the values V=7.7, corresponding

calculation for A^* reveals A^* = 0.3755, very close t_o the critical criterion. The corresponding U value is U = 1.08V = 8.32, again _somewhat later than actual instability

onset (U =6.5) of the taut strip model, as seen in fig. 1(b)

.

33
Essentially the process used here amounts to comparing two experimental performances: one, the rigid block section model

of reference 5, considered as a source of flutter derivative data, the other being the

3D taut strip model. The intermediary

employed is the theory given in eq. (1). A number of discrepancies can obviously occur between predicted and measured taut strip results for experimental and other reasons.

The process does, however, continually yield

reasonable values qualitatively consistent

with observation.

Case B/D = 2 Inspection of the source curves (Fig.

6 ) reveals potential instability regions under smooth flow as follows

2lVl3

4.7<V;^8

9.8j^V<15{?) and under turbnlent flow as follows

5.5<V6.3

9.4^V<? The criterion for flutter, eq. (4), is first calculated as follows, using data from Tables 1 and 2; = 0.0628

Employing, in the present case, B/D = 2, the process reviewed above for B/D = 1 , the values for smooth flow yield A^* = 0.0693, close to the criticality criterion, and U =6 . 6 . Reference to Fig. 2(b) reveals a sirong critical region (probably assoc^iated with vortex shedding) in the region of U - 5

,

both for smooth and turbulent flow. Subsequently, regions o^ taut-strip instability are approached in U^-14. In both smooth and turbulent flow, instability is indicated

in reference 5 for V^IO, which corresponds to U ^14; this is in fact borne out in Fig.

2^b) . Case B/D = 5 Background for this case is provided in the curves of Fig. 7. The flutter criterion

of eq. ( 4 ) in this case, with data from Tables 1 and 2, is

Inspection_of Fig. 7 for smooth flow reveals

that at V = 24.3, C^

0 j= 8.5, the value of reaches th^ above critical value. The corresponding value is

U = 1.43V = 34.75

r

Fig._ 3(b) shows in fact that in the region

of = 35 there is a strong instability

under smooth flow. Under turbulent flow the instability is postponed to U =40. However, the and corresponding A

2 * values for turbulent flow given by reference 5 (Fig. 7) do not reveal any instability tendency in the range of 0^V*50. Beyond V^O, however, they approach

zero. It is a fact (confirmed, for example,

by the curves of the case B/D = 1) that tur- bulence can increase instability tendencies.

It may be, then, that the particular intensity

and scale of turbulence used with the taut strip model differed enough from those of reference 5 that for B/D = 5, destabilizing

conditions, (i.e., necessary turbulence lev- els), missing in the tests of reference 5, were provided under the conditions of taut strip testing.

In the present case then, smooth-flow

instability of the taut strip is predicted

with the results of reference 5 but the tur- bulent flow instability is not, most likely

due to differences in turbulence level. Treatment of the Case B/D = 10 Up to this point, no background data have been introduced that are directly appli- cable to this case, to which, unfortunately,

the information of reference 5 does not

extend. An approach is suggested here, however, that permits some corroboration. During the research done for reference

3, a number of curves of A

2 * plotted vs. U/nB were acquired for various shapes of bridge deck. In particular, for a certain

set of decks, the data of Fig. 8 were assembled. For decks 2 and 3 of this group, the associated curves 2 or 3 for A* were found

to have some relation to the bluff-body case

B/D = 10. First, the flutter criterion for the taut strip (cf. Tables 1 and 2) is calculated

to be

( V)cr = while the critical reduced velocity parameter

for instability (from Fig. 4b) is U =106, which corresponds to V = 106/1.45 = 73.5 and

to K = 2 tt(B/D)/V = 0.8595 or U/nB = 7.31. Correspondingly, curves 2 or 3 of Fig. 8 yield U/nB=7 ,_f or 0.06. The corresponding value of U is U = 102.

r r It should be commented that the curves 1 to 4 of Fig. 8 were obtained in the early 1970 's by methods not necessarily as detailed as those used in reference 5 or presently employed in other studies to get A* results.

In particular, the lower (negative) values of
A^* have subsequently, in other cases, been

found to be more irregular than shown in Fig.

8, corresponding to a variety of flow

nuances, including low speed vortex excitation. In spite of this possible discrepancy,

the use of curves 2 or 3 from Fig. 8 seems

to bear out typical trends of A^* curves in the higher U/nB region of instability.

2 . 2 Summary on Prediction of Critical Reduced

Velocities for Torsional Instability

Theory, dependent on measured flutter derivatives, was used above to predict the critical

value of U as observed in the experimental

studies o5 reference 1. Table 4 summarizes

the obtainable results. Discrepancies that occur can be ascribed to a number of sources, including all measured

physical parameters. Further, the theory used is considered reliable for small amplitudes only and for spanwise uniform flow, either of which conditions could have been violated in references 1 and 5. Lastly, it

is to be emphasized that the theory applied

to this point is intended to predict critical wind velocities only, and not response amplitudes. Typically, when theoretically

critical velocities are reached, the systems

in question do not exhibit "instant" onset of "infinite" amplitudes. This is due in part

to the fact that the systems are not strictly

linear, with parameters that remain unchanged

by amplitude, frequency, etc. Also, the cross-sections dealt with are bluff, giving

rise to at least "signature" turbulence

effects, even under laminar onset flow.

3 . 1 Galloping Effects

The Den Hartog Criterion

0

( 9 ) governs incipient across-wind galloping ten- dencies in the vertical mode. (It is understood that and are both based on the same reference length.) Inspection of laminar flow lift and drag information presented in reference 8 reveals that the model with side ratio B/D = 1 is susceptible

to this type of galloping instability through

the limited range-12 < a < 12 degrees for which

dC,

< < -51 -'>»« 1.7-C„.2.3 Hence, at least in this range of ct

, the Den Hartog criterion for incipient galloping is satisfied, in fact for all wind speeds. Galloping may exhibit the nonlinear characteristic requiring motion to be initiated, however, before it can proceed. Furthermore, the Den Hartog condition is necessary, not a sufficient condition, the sufficient condition being [9];

dC, t l/2pUB(-5^t Cp)<0

( 10 ) where m is model mass per unit span and is circular frequency in cross-wind oscillation.

Case B/D=l

3. GALLOPING THEORY Across-wind galloping of a structure can be described [9] by the equation

n>[h + = -l/2oU^B + C^] J

(8) where

m = structure mass per unit span

h = vertical (across-wind) amplitude

= vertical damping ratio

oj = 2Trn^ = natural frequency of across-wind oscillation

U = wind velocity

B = along-wind dimension of object

= dimensionless lift coefficient

(referred to dimension B) a = wind angle of attac)<

= dimensionless drag coefficient

(referred to dimension B) (Note: If Cq is not referred to B, but to the across-wind dimensiog D, then

it is replaced above by ^

Talcing = -4.06, Cg^l.9, yields, for the taut strip model with B/D = 1, (see data of Tables 1,2):

U = 0.6113 m/s

This corresponds to a reduced velocity

U =1.99

r Fig. la shows in fact that a slight instability tendency in ver_tical motion under smooth

flow occurs for 1.4^ il.8.

r On the other hand, it is observed that

the presence of turbulence suppresses this galloping tendency. Therefor^, instabilities encountered elsewhere in the range are not associated with classic Den Hartog across-flow galloping for B/D=l.

It is important to emphasize that the above galloping calculation is quite sensitive to assumed values of C

‘ and . What

actual effective values ^eld during the experiment is not presently )cnown.

35
Case B/D = 2

Reference 8 provides graphs of drag, lift and moment coefficients for rectangular

sections. From these graphs, for B/D =2 ( d/b

= 0.5), values of Cq ,

, and for smooth

flow are obtainable for 0ia5:90° . Near a =0° , the slope C-j^' of the lift curve is negative,

suggesting the possibility of instability in this region. In fact

. dC - C, = = -4.73 per radian

L Qa while

These yield

-3.98

Using these values in the galloping criterion

with data from Tables 1 and ^ yields U = 0,342 m/s, corresponding_ to U = 0.93. Clearly this low value of U is noi witnessed

experimentally

.

There are some possibilities, however,

that the circumstances of the test (cf. results Fig. 2) did not correspond to the parameters used above. One strong possibility is a mean twist of the model by a few degrees, about which oscillation occurred. Another possibility is that the value of mechanical damping ratio given in Table 2 did not hold during the test wherein high

amplitude response was evidenced.

It may be observed from reference 8 that

the slope C^‘ of the lift curve for B/D = 2 makes a rapid reversal, over a very few degrees of angle of attack, from negative to positive values; this takes place in the vicinity of a =8°. If the equilibrium angle

of the taut strip model were to approach this

value, C^ could be radically different from

its a =0*^ value. Furthermore, effective mechanical damping might increase to a con- siderably greater value than that shown in Table 2. Leaving mechanical damping at its measured value, however, it is of interest to see what value of C would correspond, for example, to the ^served peak response region

U =5. This implies a velocity U = 5ri2.31) (0.03) = 1.846 m/s. Using this, with Cq=0.7, in the galloping criterion

yields the value C^= -1.438.

Again, for a doubled value of mechanical

damping (4^=0.0026), the value =-2.17 is obtained. These exploratory results tend to illustrate that quite reasonable values of damping

and lift slope could correspond to the observed experimental response seen at U =5

Fig. 2. It would experimentally be dif- ficult to measure either of these values under wind-on, oscillatory conditions.

It may be concluded that the so-called galloping theory, while encountering the dif- ficulty of correct parameter assessment, nonetheless approximates certain of the phe- nomena viewed experimentally. These phenomena, for B/D =1,2 may therefore, with

some assurance, be classified as "galloping,"

Cases B/D = 5 and 10 From the results of Figs. 3 and 4, it would appear reasonable to rule out galloping

in these cases and explore only flutter and buffeting excitations for them.

It may be remarked in general that, in the absence of detailed flow visualization

and other more deeply investigative methods

in regard to a flow-induced oscillation, it

is difficult to categorize any given phe- nomenon uniquely as "vortex-induced,"

"galloping," or "flutter," due in part to the inherently fuzzy boundaries (and, in fact, inherent shortcomings) of these classes of description

.

4. VORTEX SHEDDING EFFECTS Referring to response amplitude, those

cases investigated in reference 1 that sug- gest the vortex shedding phenomenon in strongest fashion are shown in Figs. 1 and

2, for sections with B/D = 1 and 2, respectively. When vortex-induced motion occurs it can evidence itself both across-wind and in twist. Case B/D = 1 Figure 1 shows strong displacement in both bending and torsion at the value

where n is the vertical (across-wind) fre- quency. This correspond to a Strouhal number

S = 0.15 Most researchers, however, of which reference

6 is representative, place the Strouhal

number of a square section at zero angle of attack at S=0.12 to 0,125. Fig. 9, which is

a reproduction of Fig. 3 of reference 6, sug- gests (cf, label "BT" ) that both bending and torsion excitations are associated with this Strouhal number.

36
It is o^£ interest to note that, for B/D

= 1, the value corresponding to S = 0.12

is U = 8.33, much closer to tjie first theorStically "expected" value of U (i.e., U^8.6) discussed earlier in this report

{under torsional instability). This value

was for a nominal angle of attac)c a =0°. However, in case the taut strip model twisted

("blew back") to a non-zero mean angle, this could account for a different Strouhal number holding in the experiments of reference 1. In fact. Fig. 9 suggests that a Strouhal

number S^O.15 could obtain for a mean wind

angle of about 0^12°. Another possibility is that oscillation amplitudes influenced the net effective Strouhal number. Either of these effects could explain the anomaly observed for B/D = 1. Case B/D = 2 Figure 2 suggests the possibility ^f a vortex-shedding instability close to U = 5, or S = 0.2. Again, according to Fig. § of the present report, this is far from a zero-angle-of-attack Strouhal number for B/D =2. In fact, an appropriate Strouhal number

at o=0°for B/D = 2 is about S = 0.08 (cf. Fig. 9 [6]). It is probably more nearly

correct, therefore, to assign the early peaks

(in both vertical motion and twist) of Fig.

2 to galloping than to vortex shedding.

For B/D = 5 or 10, no strong vortex-induced instability appears to be present, according to the results given in Figs. 3 and 4.

5. BUFFETING BY TURBULENCE

The theory and measurements discussed

above are concerned only with locating the critical velocities associated with instability. These theories do not attempt to predict amplitudes of response. Attempts to predict buffeting response amplitudes have

been made in references 4 and 7. The methods

of reference 4 have been employed to predict

response amplitudes for the taut strip models described herein [2]. Highlights of this work will be the subject of a future paper.

6. COMMENTS AND CONCLUSIONS

A set of experimental results presented in reference 1 has been examined. They represent the wind-induced response of a series of four taut-strip models with bluff, rectangular sections. Theories of flutter, vortex response and buffeting have been recalled from the literature and used in this report to predict the velocities of critical

instabilities of the models and the model

response amplitudes at selected points.

The general process has demonstrated that the theories used are capable of yielding results consistent with observation. However, the process of attempting to closely match experimental results reveals both the high

sensitivity of the theories to certain parameter inputs and the need to acquire highly accurate parameter estimates in order to obtain dependable results from the predictions . The data provided by reference 1, while

extensive, were not fully adequate to support

the theories used. Neither were the literature sources relied upon in the present

study

.

Clearly, extensive further work could be done

in this area to fill the gap in knowledge

required to tighten up the effectiveness of the theoretical predictions. In particular, whole sets of accurate flutter derivatives

were lacking for the present study, forcing recourse to somewhat inadequate literature

data and to the pressing into use of static derivative data as a substitute for true flutter derivative information. In spite of these difficulties, the theories called up appeared to yield plausible results based on reasonably argued parameter choices. Generally, experiments with taut strip models should be accompanied by rigid, sprung sec- tion model experiments for the same cross-sections. Such models are accurate

sources of indispensable data to be used in forecasting taut strip behavior. The theories used were found most sensitive to the input parameters concerned with aerodynamic

derivatives, both static and dynamic, and to damping values. REFERENCES

1. Marshall, R.D., Yamamoto, K., Nakagami

,

Y., and Kanjaki, K., "Response of Taut Strip Models to Smooth and Turbulent Flows," Proceedings of the Fifteenth Joint UJNR Panel Conference on Wind and Seismic Effects, Tsukuba,

Japan, May 1983.

2. Scanlan, R.H., "Comparison of Theoretical and Experimental Results In Response of Taut Strip Models to Smooth

and Turbulent Flows", Report prepared

for National Bureau of Standards, August

1985.

3. Scanlan, R.H. and Tomko, J.J., "Airfoil and Bridge Deck Flutter Derivatives,"

Jnl. Enqrq. Mech. Div. , ASCE, Vol . 97, No. EM6, Dec. 1971, pp . 1717-1737.

37
4 . Scanlan, R,H., "State-of-the-Art Methods

for Calculating Flutter, Vortex-Induced,

and Buffeting Response of Bridge Structures," Report FHWA/RD-80/050

,

Federal Highway Administration,

Washington, D.C. , April 1981.

5. Nakamura, Y. and Yoshimura, T., "Flutter

and Vortex Excitation of Rectangular

Prisms in Pure Torsion in Smooth and Turbulent Flows," Jnl. of Sound and Vibration (1982), Vol . 84, No. 3, pp. 305-317.

6. Otsuki, Y. , Washizu, H., Tomizawa, H., Ohya, A. and Fujii, K., "Experiments on the Aeroeslastic Instability of Prismatic Bars with Rectangular

Sections," Proceedings, Conf. on Wind

Effects on Bldgs, & Structures, Tokyo,

1971, Paper IV 24, pp. 891-898

7. Scanlan, R.H, and Gade, R.H., "Motion of Suspended Bridge Spans under Gusty

Wind," Jnl. Struct. Div. ASCE, Vol. 103,

No. ST9, Sept. 1977, pp. 1867-1883.

8. Okauchi, I., Ito, M. and Miyata, T. , Wind

Resistant Structures (in Japanese),

Tokyo, 1977,

9. Simiu, E. and Scanlan, R.H., Wind

Effects on Structures , Second Edition, John Wiley & Sons, New York, 1986.
Table 1

.] PHYSICAL PROPERTIES OF TAUT STRIP MODELS

B/D ^T H

o P B D L Dimension (Kgf m (Kgf m s^) (Kgf m s (m) ( m) ( m)

1 0.0437 0.2625 0 . 12 0.06 0.06 1 . 3 2 0.0317 0 . 1244 0.12 0 .06 0.03 1 . 3 5 0.0240 0.0821 0 .12 0 . 06 0.012 1 . 3 10 0.0153 0.0535 0. 12 0.06 0.006 1 . 3

B/D

1

2

5 10

Table 2 MECHANICAL DAMPING RATIOS AND NATURAL

FREQUENCIES OF TAUT STRIP MODELS

B/D ^h Dimension (Hz .) (Hz . ) "h=^

1 0.0011 0.014 10 . 23 11.05 1.08 1 .08

2 0.0013 0.0051 12 . 31 16 . 74 1 . 36 2 . 72

5 0.0010 0 .012 13. 98 19.99 1.43 7 . 15

10 0.0018 0.011 15. 86 23.00 1 .45 14 . 50

Table 3 CHARACTERISTICS OF TURBULENT BOUNDARY LAYER

Z

( cm) (Percent) (Percent)

Lu

( m) Lw

(m) 35 0.62 20 .

8

14 .

7

0.21 0.08

60 0 . 85 11.4 8.8 0 .22 0.07

70 0.91 7. 1 6 . 2 0 . 20 0.07

Table 4 COMPARISON OF CRITICAL VALUES OF U IN

r TORSION

Smooth Flow Turbulent Flow Theory Taut Strip Experiment Theory Taut Strip Exper

8.6(?) 6 .

6

8 . 3

6 . 6 6.8;14 ~5 .0;~14.5 -4 . 8 ; 14 ~5 . 0 ; >14 34.8 -36 40 102 106 - 85

39
RMS

TORSICmL

DISPLACmn'T’

(decj)

RMS

VERTICAL

DISP]

1 -1 B/D = 1 iw = 14.7 X

(a)

X

a . 8.8

= 6.2

- SMOOTH FLOU -

- X

-

B

V

- o; - 0

s ftad,.

X

p. X *0 ° 0 , 1

j 1 2 3

in 8 S

B/D = 1

-1 1 1

! lu = 14.7 X

-

(b) X = 8.8

= 6.2

0 SMOOTH FLOU

- -

- -

X

“

f

t);%Pa''b;px 0- ick q|> X^B

X A

, -X

, V ° V , ,

[

I

Fig. 1. RMS Displacement at Midspan. B/D = 1.
RMS

TORSIONAL

DISPLACEMENT

(deg)

rMS

VERTICAL

DISPLACEMENT

(rrni)

B/D = 2

0

0

• lu = 14.7 X

X - 8.8

0 = 6.2

0 SMOOTH FLOU

(a)

0

JBi |Q:> > I

V

V' I 8 10 12 14 16 18

Fig. 2. RMS Displacement at Midspan. B/D = 2.

41
PflS

TORSiaSIAL

DISPIACTMTvIT

(deg)

RMS

VERTICAL

DISPLACEMENT

(inn)

5

3 -

1 -

» Iw = 14.7 X

X = 8.8

0 r 6,2

SMOOTH FLOU

(a)

0

.

-

" "x

x"a ^ ^ C ^

10 15 20 25 30 35 40 45

B/D = 5 U - 14.7 X

X = 8.8

: 6.2

0 SMOOTH nou

(b)

2 -

1 -

X <P

X

1^

°C °0| 0 0| ^

10 15 20 25 30 35 40 45

Fig. 3. RKS Displacement at Midspan. B/D = 5.

42
RMS

TORSICmL

DISPLACEMENT

(deg)

RMS

VERTICAL

DISPLACEMENT

(nm)

5 B/D = 10 Iw = 14.7 X

X = 8.8

= 6.2

SMOOTH FLOW

(a)

1 r

0 90 100 110

1 -

« . fa' j, % ^ , 10 20 30 40 50 60 70 80 90

B/D = 10 lu = 14.7 X

= 8.8

= 6.2 SMOOTH FLOU

(b)

1 r

s <>0 90 100 110

•a #

0 10 20 30 40 50 60 70 80 90

Fig. 4. RMS Displacement at Midspan. B/D = 10.

43
2-5

Fig. 5

Fig . 6

Fig. 7

Torsional Damping Coefficient. B/D = 1, (Ref. 5)

° Smooth Flow

• Turbulent Flow

Torsional Damping Coefficient. B/D = 2. (Ref. 5)

Torsional Damping Coefficient. B/D = 5. (Ref. 5)

44
Strouhal

Number

Fig. 8. Aerodynamic Coefficients. (Ref. 4)

"B" denotes Strouhal numbers closely related

to critical speeds for across-wind

oscillation.

"T" denotes Strouhal numbers closely related

to critical speeds for. torsional

oscillation.

Fig. 9. Strouhal Number in the Wake of Rectangular Sections.

(Ref. 6)

45
A FULL-SCALE MEASUREMENT OF WIND LOAD ON A SOLAR WATER HEATER

BY Hisashi Okada and Tatsuo Murota^

ABSTRACT

Results of a full-scale field observation of wind loads on a solar water heater, mounted on a house roof, are described and compared with wind

tunnel tests obtained by the authors. Wind

force coefficients estimated from wind tunnel

tests were 10 to 20% larger than those observed

in the field. This suggests that the design

pressure coefficients obtained by wind tunnel

tests are good estimates for designing solar heater mountings.

1. INTRODUCTION

Solar water heaters for energy conservation are coming into wide use in Japan. Most are mounted on roofs of residential houses where

they are exposed to strong winds. It is the structural engineer's concern to prevent heater

damage from high winds. It is difficult to assess the possibility of damage because

technical information is scarce. Under those circumstances, the authors conducted wind tunnel

tests(l) in 1982 to produce information about wind forces on solar water heaters. Full-scale

field observations were performed at the same

time. This report presents the outline of the field observation and compares wind forces obtained by wind tunnel tests and field observations.

2. DESCRIPTION OF FULL-SCALE OBSERVATION

2 .1 Observation Field

A full-scale test specimen was located in a flat open ground at the south east corner of the site of the Building Research Institute, Ministry of Construction (Fig. 1 and Photo 1). A transmission tower is located about 30m southeast of the specimen. Except for the tower, nothing else is within the area 40m to the west, 70m to the east, and more than lOOm to the north

and south of the specimen.

2 .2 Test Specimen

Test specimen is a full-scale solar water heater mounted on the center of a roof as shown in Fig 2. The house is 9m x 5.46m in plan and has

a gable roof of 3/10 pitch. The height of the eaves is 2.54m. The heater has a laid-L-like

section, and its size is 2m x 2.23m in plan. It was mounted on a steel frame fixed to the house

roof of 30° inclination to the horizontal plane. Three sets of load cell were installed between

the heater and the steel frame to measure wind

forces on the heater.

Upper surfaces (collector) of solar water heaters generally face south. In this observation, however, it faced east taking predominant direction

of winds into consideration. Therefore, in applying the observation results, wind directions should be changed by 90 degrees: east and west for observation must be considered to be south and north, respectively.

2.3 Wind Direction and Wind Speed Measurements

To observe wind direction and wind speed in the field, a 10m high mast was erected 6m southeast of the specimen as shown in Fig. 1. A vane-type anemometer and three 3-cup-anemometers

were mounted at the top, 6, 4 and 2m levels of the mast, respectively.

2 .4 Data Acquisition

Wind and load data observed were processed by

the following two methods: 1) data are sampled

at 20 second intervals during 10 minutes at every 1 hour and the 10 minutes are recorded,

2) data are sampled and recorded at 1 second intervals during 10 minutes. The first method

was adopted for strong wind conditions. Turbulence intensities and power spectra presented

later are calculated using the data acquired by

the second method.

3.

WIND TUNNEL TEST

Before presenting the results of field observation, the wind tunnel test in 1982 will be described.

3.1 Model

Fig. 3 shows a model for the wind tunnel test. The size reduction ratio is 1/20. The house with the heater mounted on its roof is assumed

to be 2 stories and with plan dimension of 6.4m

X 8.2m. It has a gable roof of 3/10 pitch. A heater is assumed to be a flat plate type having

the size of 2m x 2m in plan. It is mounted on a ridge in 30° inclination to the horizontal plane.

3.2 Wind Profile in Wind Tunnel Test Profiles of mean wind velocity and turbulence intensity are shown in Fig. 4 and Fig. 5, respectively. Mean wind velocity profile

‘^Building Research Institute, Ministry of Construction, Tsukuba
approximated to the following log-law equation: U(Z)c< Ln(Z/Za) (1) in which, U(Z) = mean wind velocity (m/s) at height over wind tunnel floor

Z(m)

Zo = roughness parameter

= 0.70cm(14cm in conversion to full-scale) Turbulence intensity at ridge height is about

15% and a power spectrum of turbulence at the same height is shown in Fig. 6. It approximated

the Davenport's empirical equation(2):

+ ( 2 ) a oQ

velocity profiles of east and west wind at rather

strong velocities, respectively. Here, directions west or east correspond to the true west or east jt 12.5°. Fitting these profiles for eq. (1), we get roughness length Zo of about 5cm. This is considerably small compared with wind tunnel in open fields. Fig. 10 shows a turbulence intensity profile in west winds. Turbulence intensities are 30 to 40% and are greater

than those for wind tunnel flows. Fig. 11 shows power spectrum of turbulence at 10m high in west winds. It approximates Davenport's spectrum

for X=200n as shown in the figure. The wave length giving a peak value of power spectrum

agrees with one proposed by Davenport for natural

winds, but it is much larger than that for wind

tunnel flows.

4 .2 Wind Forces

in which, S(n) = power spectral density at n(Hz)

= variance of turbulence

X = nL/U(10)

L = n/Tx wave length giving power

spectral peaks

U(10) = mean wind velocity(m/s) at height of 10m in conversion to full-scale

Sprectrura in Fig. 6 was expressed by eq. (2) with L=14.6m (U( 10)=8 . Im) . The value of L was converted to 298m in full-scale.

3.3 Summary of Results

Relationships between 10 minute mean wind forces

and 10 minute mean wind velocities at 10m high

are shown in Figs. 12 and 13 for west and east

wind, respectively. Positive and negative signs

of wind forces are used following the same defi- nition as stated in Fig. 7. These figures show

that wind forces on heaters increase in proportion to the square of wind velocity. Curves of quadratic equation in these figures show regres- sion equations obtained by assuming the rela- tionship between wind forces(F) and wind velocities(UdO) ) at 10m high to be

F = a U(10)^, where a is a constant. Wind force coefficients can be derived from the constants

as follows: Values of wind force coefficients for the heaters

changed with wind directions as shown in Fig. 7. Those coefficients are normalized by the dynamic

pressure at height of eaves as follows:

C = F/A/qo (3) in which, C

FA

qo

= wind force coefficient

= mean wind force

= projected area of heaters to a plane parallel to lower surfaces

= mean dynamic pressure at eaves height Positive and negative signs of coefficients are defined as: when wind forces push heaters toward

the roof, corresponding coefficients are positive

and when wind forces pull heaters off roofs, corresponding coefficients are negative. Fig. 7 shows wind force coefficients are positive in directions 0°to 90°, and are negative in directions 90° to 180°. The values of coefficients

for 0° and 180° are about 0.80 and -0.97, respectively. The coefficients reach a positive

peak value at about 45° and a negative peak value at about 135°.

4. RESULTS OF FIELD OBSERVATION

4.1 Characteristics of Wind

Wind directions observed are mostly west and

east. Figs. 8 and 9 show 10 minute mean wind

1 , <1 < a

in which, p

= air density

Uj.gf = 10 min. mean wind velocity

at height of eaves

C is calculated as -0.87 or 0.57 in case of west

or east wind, respectively.

5. DISCUSSION OF RESULTS

Values of wind force coefficients and corres- ponding major parameters for full-scale field observations listed in Table 1 are compared with

those for wind tunnel tests. Values of wind

force coefficients observed outdoors are 10 to 20% less than those obtained in the wind tunnel. The differences are caused by the effects of:

(1) Scale of turbulence

(2) Intensity of turbulence

(3) Re 3molds Number

(4) Sectional figure of heater

(5) Number of house stories

It was demonstrated by wind tunnel tests that

the effects of parameters (4) and (5) of wind

forces on heaters are small. Parameter (3) has little effects of wind forces on bluff bodies. Therefore, differences in values of parameters

(1) and/or (2) in wind tunnel tests and full- scale field observations are considered major

47
reasons that caused the difference in wind force coefficients

.

6. CONCLUDING REMARKS

Wind force coefficents obtained by wind tunnel

tests are 10 to 20% larger than those from fullscale observations. Therefore, by adopting wind

tunnel test results to design of heater mountings, wind loads will be estimated with a factor

of safety. To determine a more accurate estimation of wind forces on solar water heaters,

effects of scale and/or intensity of turbulence

should be investigated.

7, REFERENCE

(1) Hisashi Okada and Tatsuo Murota, "Wind Loads

on Solar Water Heaters," 14th Joint Meeting,

U.S. -Japan Panel on Wind and Seismic

Effects, 1982.

(2) A.G. Davenport, "The Dependence of Wind Loads on Meteorological Parameters," Wind Effects on Buildings and Structures, 1967.
Table 1 Comparison between wind tunnel test and full-scale observation

wind tunnel test full-scale observation wind force coefficient

(west wind) -0.97 -0.87

(east wind) 0.80 0.57 wave length giving

a power spectral peak 170m 660m

turbulence intensity

at height of eaves 15/^ 35^ roughness parameter lUcm 5cm plan and dimensions rectangle rectangle

of heater 2m X 2m 2m X 2.36m cross-sectional fig-ure flat plate laid-L-like

of heater

plan and dimensions rectangle rectangle

of house 6 . Um X 8.2m 5.^6m X 9m number of story of house 2 1 roof gable roof gable roof

pitch of roof 3/10 3/10
Fig.l Layout of observation field

Photo 1 Observation field viewed from W-N-W

50
]

]

]

]

]

]

]

]

“1J

]

]

]

]

]

]

]

]

]

Fig. 2 Specimen for field observation

uu)/uuoo) Fig.U Mean wind velocity profile

of wind tunnel flow

Fig. 3 Model for wind tunnel test

O

<5

(

^

1 S °o N o $

0 10 20 30 40 4Z2/U (’/.) Fig. 5 Turbulence intensity profile

of wind tunnel flow

51

ooe
NORMALIZED

POWER

SPECTRAL

DENSITY

Fig. 6 Power spectrum of turbiilence of wind tunnel flow Fig.T Wind force coefficients

(wind tunnel tests)

Fig. 8 Mean wind velocity profile

(west wind)

Fig. 9 Mean wind velocity profile

(east wind)

52
Fig. 10 Turbulence intensity profile Fig. 11 Power spectrum of turbulence

(west wind) (west wind)

Fig. 12 Relationship of mean wind force to wind velocity (west wind)

U(I0 ) mit Fig. 13 Relationship of mean wind force to wind velocity (east wind)

53
DESIGN CRITERIA FOR GLASS CLADDING SUBJECTED TO WIND LOADS

by

Emil Simiu^ and Erik M. Hendrickson^ ABSTRACT The safety and cost of glass cladding facades snbjected to wind loads depend upon the criteria

used for their design. These criteria specify the design wind loads and the allowable panel loads.

The present paper reviews: the evolution in recent years of methods for the estimation and

specification of design wind loads for building facades; recent developments concerning the estimation and specification of allowable loads for glass cladding panels, and controversies surrounding these developments; and recent

developments in structural reliability as applied

to glass cladding, which allow rational decisions

to be made concerning the adoption of satisfactory

design criteria specifying both design wind loads

and allowable panel loads. It is shown that, in view of recent changes in methods for estimating

design wind loads and allowable panel loads, it would be imprudent or wasteful to adopt a standard specifying allowable panel loads without giving due consideration to the overall reliability aspects

of the cladding glass design problem. Comments are offered concerning the development of such a standard. The paper is addressed to members of the ASTM E06.51.13 Task Group on the Structural

Performance of Glass and to other individuals and bodies concerned with the behavior and design of glass cladding under wind loads. KEYWORDS

Design; Glass cladding; Probability; Wind Loads

.

1. INTRODUCTION

A glass panel subjected to wind loads is designed

so that the design wind load , q^, does not exceed

the design value of the load capacity, q^, referred

to as the allowable load . A design criterion

consists of the specification of the values q^ and

‘’a* To each set of values q^ and q^^ there correspond probabilities P^ and P^, representing,

respectively, the probability that the wind load

on the panel will exceed the value q^ in any one year and the probability that the panel will break

if subjected to a load q^. A design criterion may thus be redefined as consisting of the

specification of the probabilities P^ and P^^. For example, it might be specified that panels be designed so that P^ = 0.02/year (corresponding to

a 50-year mean recurrence interval of the wind

load), and P^ = 0.008 (corresponding to a load that, if applied to the panel, would cause eight

out of 1,000 panels to fail). It is required that

the specification of q^ and q^ (or of P^ and P^^) be such that the safety level for the cladding

glass facade as a whole is acceptable. It is noted

that the safety level depends on both q^ and q^^ (or P^ and Pj^) . Thus, assuming that a design criterion has been proven by experience to result in acceptable designs, such a criterion should not be tampered with by changing the specification of the wind load without modifying accordingly the specification of the allowable load, and conversely.

The purpose of this paper is to examine various proposed design criteria in light of this observation. Such an examination is rendered necessary by various changes in the specification

of design wind loads q^, or of allowable loads q^

,

proposed in the last decade without sufficient regard for the inextricable interconnection of

these two quantities from a reliability viewpoint. To provide the requisite background for this effort, we first review briefly the evolution in recent years of methods for the estimation and specification of design wind loads for building facades. We then review recent developments

concerning the estimation and specification of allowable loads on glass cladding panels, and controversies surrounding these developments. Finally, we review the structural reliability

framework that enables rational decisions to be made concerning the adoption of a design criterion

specifying both design wind loads and allowable

panel loads.

In attempting to assess various proposed design

criteria, we will define 'acceptable safety level’

as one that, in light of past experience with the behavior of glass facades under wind loads, is satisfactory from a public safety point of view

and does not entail wasteful overdesigns.

2. DESIGN WIND LOADS

It will be useful for our purposes to illustrate

two methods that have been widely used to specify

design wind loads. To this end we reproduce two examples of wind load calculations from the Technical Service Memorandum issued by the Glass Division, PPG Industries, titled 'Design Wind

Loads, ANSI A58. 1-1972,' dated June 21, 1974. The examples show the calculation of wind loads in accordance with the City of Atlanta Building Code

and the American National Standard ANSI A58. 1-1972. They pertain to a 12-story building with dimensions

50 ft by 50 ft (15.24 m x 15.24 m) in plan and 150

ft (45.72 m) tall, assumed to be located in downtown Atlanta, Georgia.

^Center for Building Technology, National Bureau

of Standards, Gaithersburg, Md

.

54
2.1 City of Atlanta Pnilding Code Reonirements The City of Atlanta Building Code wind load requirements, as quoted in the PPG Memorandum, are listed below. Peight of Structure Wind Pressure

(feet) (psf) Under 30 25

30 - 49 30

50 - 99 40 100 - 499 45

500 - 1199 55 1200 and Over 60

(1 ft = 0.3048m; Ipsf = 47.9 kPa)

The design pressure distribution required by the City of Atlanta Building Code is represented in Fig. 1, extracted from the PPG Memorandum. Note

that the pressures at any one elevation of the building are uniform, i.e., pressures on corner and center panels are assumed to have the same value.

2 .2 ANSI AS8. 1-1972 Peauirements

Figure 2, also excerpted from the PPG Memorandum,

shows pressures and suctions on the glass panels

of the building, calculated in accordance with the provisions of the American National Standard ANSI

A58. 1-1972 [1] under the following assumptions:

(1) the building is not exceptional in terms

of its degree of sensitivity to wind or of its degree of hazard to life and property, and

(2) the terrain upwind of the building is built up for at least one-half mile in all directions, with at least 50 percent of the buildings in excess of six stories; in addition,

the terrain for at least one more mile is of the suburban type.

In Fig. 2, positive and negative numbers indicate pressures and suctions, respectively. It is seen

that pressures on corner strips (having, according

to the ANSI Standard, width = 0.1 x 50 ft = 5 ft (1.52 m) on all faces) are higher than pressures

elsewhere on the building faces. (The pressures

in Fig. 2 are consistent with the provisions of

ANSI A58. 1-1972 except for a few minor numerical

differences.

)

Three points concerning the ANSI Standard

requirements deserve comment. First, these requirements make allowance for the effect of terrain exposure on the wind loads. In contrast,

the City of Atlanta Building Code, and similar

codes, specify the same wind loads regardless of whether the building being designed benefits from

a favorable exposure or not. It is seen from Figs.

1 and 2 that the maximum (corner) loads specified

by the ANSI Standard are in this case lower than

the loads specified by the City of Atlanta Building

Code. However, it can be verified that they would

have been higher had the conditions listed under

assumption (2) above not been satisfied.

Second, the requirements of the City of Atlanta

Code (and of similar codes) imply the use of a single pressure coefficient for the entire

building. In contrast, the ANSI Standard specifies

a different pressure coefficient for each distinct

aerodynamic zone of the facade. For example, the negative external pressure coefficient at a building comer is C^ = -2.0. Thus, as seen in Fig. 2 and pointed out earlier, loads are not uniform over the width of the building, and if architectural and construction considerations

permit, it is possible to limit the use of stronger

glass to the zones of high pressure, rather than using the same glass thickness throughout the entire facade. Third, in the ANSI Standard — as in the City of Atlanta Building Code Loads — the directional

features of the extreme wind climate and of the aerodynamic flow around the building, are not accounted for. Indeed, the external pressures calculated in accordance with the ANSI Standard are proportional to the product of the appropriate pressure coefficient and the square of the wind speed from any direction . Vj^, corresponding to a N-year mean recurrence interval (usually, N = 50, or, in the case of buildings with an unusually high

degree of sensitivity to wind, or an unusually high degree of hazard to life and property, N = 100). Since our object is to review only the broad

features of methods for specifying design wind

loads, we do not discuss here revisions included

in the 1982 version of ANSI A58.1. Although pressure distributions in ANSI A58. 1-1982 are somewhat different from those indicated in Fig. 2, in general they would not affect the three points

just commented upon.

2.3 Design Loads Based on Wind Tunnel Tests Methods for estimating design wind loads have evolved in recent years following the increasingly

widespread use of the wind tunnel as a design tool . In the first years during which wind tunnel testing

of individual buildings became relatively common, external pressures were calculated by using

essentially the same method as the ANSI Standard, except that pressure coefficients were determined

from wind tunnel measurements, rather than being

taken from ANSI Standard tables. In wind tunnel tests external pressure coefficients are measured

at a large number of points on the building facade

for as many as 32 or more directions. The pressure coefficient at a point was assumed to be the largest of the coefficients measured for the various wind directions at that point. The design pressure was then assumed to be proportional to that pressure coefficient and to the square of the wind speed, Vjyj. Thus, the dependence of pressures upon the directionality of the extreme wind climate

was ignored, just as in the 1972 or 1982 ANSI Standard. It is convenient to summarize the procedure for calculating design wind loads just described by the formula

q = 1/2 p max[Cp(a) Ivj^j^ (D

55
where Cp is the pressure coefficient for the point

being considered, a is the wind direction, p is the air density, and is the design wind speed.

The main motivation for condncting wind tnnnel studies was to identify areas on the building facades where pressures could be higher than indicated in tables, such as those included in the ANSI Standard. Areas with high pressures could be expected to exist on buildings with unusual geometric shapes, re-entrant comers, and so forth,

or buildings subjected to channeling or other

aerodynamic effects due to the proximity of other stmctures. It was found that areas with pressure

coefficients higher than those specified in the ANSI Standard or than those implicit in various

codes are quite common indeed. At the same time

it was found that, most commonly, high pressures

were not only confined to relatively small areas, but also corresponded to pressure coefficients that decreased rapidly as a function of direction, as suggested, for example, in Fig. 3. It could, then, be argued that the chances of the wind speed having

the value Vj^ and having precisely the most

unfavorable possible direction are small, and that

a reduction of design pressures would be in order. For reasons not explained satisfactorily, a blanket

reduction factor of about 0.8, meant to account for

the favorable effect of directionality, has been

proposed (see, e.g., [2]).

It is in fact possible to estimate wind loads with

any given mean recurrence interval by taking into account both the directional dependence of the pressure coefficients and the directional dependence of the extreme wind climate. For

details, see Bef. 3, Chapt. 8. To illustrate the effect of directionality, consider the following

simple example. Assume that the pressure

coefficient and the largest wind speed in a given

year are Cp(oj) = 1.0 and v(aj) = 30 m/s for direction , and that ^^( 02 ) = 2.0 and v(o2

) = 20 m/s for direction 02 . The air density is assumed to be p = 1 .2 kg/m^. If the pressure is obtained by an expression similar to Eq. 1, the calculated largest pressure for that year would be

q = 0.5 X 1.2 X 2.0 x 30^ = 1080 Pa Actually, at the point of interest, during the year

in question, the largest pressure from direction

is q(aj) = 0.5 X 1.2 x 1.0 i 30^ = 540 Pa

and the largest pressure from direction 02 is q(c2

) = 0.5 X 1.2 x 2.0 x 20^ = 480 Pa

i.e., the largest wind pressure during the year is 540 Pa, rather than 1080 Pa, as would be indicated by the calculation based on the largest wind speed

from any direction. In reality, we would of course

have to perform calculations for, say, 8 or 16 directions. However, from this deliberately simple example it can be inferred that, in most cases, a wind load with a 50-year (or 100-year) mean recurrence interval calculated by taking wind directionality into account is smaller than the

wind load calculated by Eq. 1 in which the wind

speed V|^j is the wind from any direction, obtained,

for example, from the ANSI Standard. Within the last decade, rather than using Eq. 1, commercial wind tunnel laboratories usually have provided information on design wind pressures based

on calculations that account for the directional

dependence of both the pressure coefficients and

the extreme wind climate. Interestingly, the mean

recurrence interval is quite often not specified

by the building official, perhaps because not enough building officials are sufficiently

familiar with the relatively new developments and issues involved in such specification. It is common

for the wind tunnel laboratory to provide design

wind pressures corresponding to 50- and ICO-year mean recurrence intervals, and the choice of which

to use in design is left to the developer or building owner, in spite of the public safety

issues such a choice obviously entails.

It is clear from what has been said above that design pressures with a 50- or 100-year mean recurrence interval calculated by taking into account the directional dependence of the pressure

coefficients and of the extreme wind climate may

be lower than the values obtained from the ANSI

Standard or from codes of the type exemplified by

the City of Atlanta Building Code. The question

then arises whether the safety levels of glass cladding facades designed on the basis of such pressures are acceptable. To attempt an answer

to this question it is first necessary to examine

the matter of estimating and specifying allowable

loads, q^.

3. ALLOWABLE LOADS ON GLASS PANELS

3.1 Design Charts Although other sources of information on this topic exist, the major documents available for use by

designers in the United States are those issued by

LOF Glass [4], PPG Industries [5,6], and the American Society for Testing and Materials (ASTM

F 330 Standard) [7]. These documents specify 60-second loads causing glass panel failure. Differences among the charts of [4], [5], and [7] are relatively minor. While [4] and [7] are essentially unmodified versions of documents that have been in existence for almost two decades, the PPG Industries charts first issued in 1979 [6] represent a major departure from the PPG Industries charts issued in 1974 [5], and therefore also differ substantially from [4] and [7].

It is useful at this point to present the following excerpts from the LOF Technical Information

document AT8-108 [8] concerning the ASTM charts [7] and the departures from them in the charts of [6]

:

"Starting in 1964 a consensus on glass

strength began to develop within the building

industry. Within five years, strength of glass information in building codes and the standards of trade associations and technical societies became uniform for practical

56
purposes. Of these documents the one exposed

to the most critical study is ASTM Standard

E 330, Structural Performance of Exterior

Windows, Curtain Walls, and Doors by Uniform

Pressure Difference. A glass strength chart

included in the appendix to this Standard is ... generally considered to be an 'industry standard' for glass exposed to uniform

loads ...

.

"On April 23, 1979, the Glass Division of PPG Industries issued a technical bulletin on

glass strength.... The information therein

differs significantly from the ASTM chart and

all other current publications....

"The PPG charts are more conservative than existing information. In some cases the discrepancies in the PPG information are large.... The specification writer faces a dilemma. His indicated choice of glass may

depend upon the manufacturer. PPG information

suggests one thickness and the data used by

the remainder of the industry will lead him

to another thickness, most often thinner....

load corresponding to a probability of breakage P^

= 0.008 as given in the LOF charts [4], the 1974 PPG Industries charts [5], the ASTM E 330 Standard

[7], the 1979 PPG charts [6], and the E06.51.13

draft, for two annealed glass panels supported on

four sides: panel 1 is assumed to have a 15 ft^ (1.39 m^ ) area, 3/16 in. (5 mm) thickness, and aspect ratio (ratio of long side to short side) equal to 2.0; panel 2 is assumed to have a 20 ft^ (1.86 m^) area, 5/16 in. (8 mm) thickness, and aspect ratio 1.0. It is noted that the dependence

of the allowable loads on aspect ratio is reflected only in the 1979 PPG charts [6] and in the charts

1 the E06.51.13 draft document,

listed below.

The loads q^

Reference Panel 1 Panel 2 LOF [4] 52 psf 82 psf PPG (1974) [5] 50 psf 78 psf

ASTM [7] 50 psf 81 psf PPG (1979) [6] 28 psf 70 psf E06.51.13 Draft 29 psf 50 psf

1 psf = 47.9 kPa

"In a letter of February 13,1979 to ASTM, PPG characterizes the ASTM chart referenced earlier as presenting 'possibly dangerous

information.' Persuasive reasons for these opinions are not included...."

The drastic changes embodied in the 1979 PPG charts

[6] might well have appeared all the more

surprising to the industry as the previous PPG

document [5] (whose charts did not, by and large, differ substantially from the LOF [4] or the ASTM

[7] charts), contained a section entitled 'Current Practice on Firm Foundation,' which stated: "For more than 20 years, empirical tables developed by PPG were accepted internationally

as references for determining glass thickness and size. A chart including failure

probability considerations, published in 1962, established a more reliable procedure, which

is now in general use. The new product

recommendations presented here are based on recent research, analysis, and experience."

Additional comparisons among values taken from various sources are presented later. Reference 4 contains information, based on test results, on allowable loads corresponding to failure probabilities P^ ranging from 0.2 to 0.0005. Similar information is given in [6] for probabilities P^ ranging from 0.5 to 0.001, based

on the assumption that the strength of glass has

a normal distribution with a coefficient of variation of 25% for annealed glass and 20% for heat strengthened glass.

3.2 Sources of Differences Among Load Charts Differences among loads corresponding to a failure probability P^ = 0.008 specified in the various

charts can be attributed to: (1) differences in the definition of the basic strength of glass as affected by in-service degradation, (2) differences

among the respective methods for estimating the load capacity, and (3) effects of statistical

errors

.

Prompted by discrepancies between [6] and other

charts, the task group ASTM E06.51.13 (Structural Performance of Glass) of the ASTM E06.51 Committee

on Component Performance of Windows, Curtain Walls

,

and Doors is currently examining a draft proposal submitted by the task group chairman, which

includes proposed charts for allowable loads q^ on annealed glass panels simply supported on four

sides, corresponding to a probability of failure

P^ = 0.008. In a number of instances, especially

for thinner glass, the loads in these proposed charts are relatively close to those of [6]. However, in other instances the loads in the proposed charts are substantially lower than the loads in [6] (and, therefore, even more

substantially lower than the loads in [4], [5], and

[7]). To illustrate differences among various

available charts, listed below are values of the

3.2.1 Effect of In-Service Degradation

The 1974 PPG document [5] states: "The curves plotted for each nominal thickness represent the expected performance of glass of PPG minimum

thickness after normal surface depreciation has occurred. These curves are not applicable to glass

that has obvious surface damage, such as abrasion, scratches, or welding splatter tracks.... In the interest of public safety, prompt replacement of damaged glass is imperative."

According to [4], the LOF charts "are based on testing glass to destruction. In effect they are

a summary of test results." It is further stated

in [4] that "float glass, plate glass, and sheet glass had the same strengths for practical

purposes. Any differences are not significant

57
enough to warrant distinguishing between the three methods of production." No explicit statements are made in [4] on whether any allowance was made for surface degradation, but from the similarity between the values of [4] and [5] one might infer that surface degradation was accounted for in essentially the same manner in the respective sets

of charts.

As noted earlier, the loads in the new PPG charts

[6] are in most cases substantially lower than

those given in the 1974 PPG document [51. The principal reason for these reductions lies in the decision by PPG to reduce the strength of annealed glass by 33 percent to account for in-service

surface degradation, presTimably beyond what was referred to in the earlier PPG document [5] as normal surface depreciation. This decision is justified as follows [9]: "While the average breaking stress of as-produced glass is approximately 9000 psi; limited strength data using in-service float glass, in conjunction with

engineering judgement and experience, indicate that

a 60-second average breaking stress of 6000 psi with a coefficient of variation of 0.22 probably

represents the strength of in-service float glass reasonably well for architectural design purposes with respect to wind loads." Note that the 'average breaking stress' of 6000 psi and the coefficient of variation of 0.22 are also listed

in [6] for annealed glass.

As far as the E06.51.13 Draft is concerned, it is the writers' understanding that it is based on

tests conducted on glass panels removed after about 20 to 25 years of service from three buildings

subjected to renovation or wrecking and salvage operations [10]. Of these buildings, the Great

Plains Life Building was severely damaged during

the May 11, 1970 Lubbock tornado, which caused the

loss of a substantial part of the glass panels

(e.g., on the south and west elevations of the building, about 50 percent of the panels were lost

[11]). The reduction in strength for in service

glass with respect to new glass given in [10] is on the order of 66 percent, rather than 33 percent

as in the case of [6].

3.2.2

Effect of Methods for Estimating Load Capacity

As mentioned above, the LOF charts were obtained

by fitting curves to test results. The charts in

[6] were developed using the following criterion:

panel breakage occurs if the maximum stress in the plate is equal to the breaking stress. It was further assumed that for annealed glass the breaking stress has a normal distribution with a mean of 6000 psi and a coefficient of variation of

0.22, as indicated earlier [9]. For example, to calculate the breaking load corresponding to a

0.

008 probability of panel failure, it is necessary

to obtain the breaking stress corresponding to the value 0.008 of the cumulative distribution function

of the breaking stress. From normal distribution

tables it follows that this breaking stress is equal to the mean breaking stress minus 2.41 times

the standard deviation of the breaking stress,

1.

e., 6000 - (2.41 x 0.22 x 6000) = 2819 psi. (The

value listed in [6] is 2800 psi.) The load that corresponds to a 0.008 failure probability for the panel being considered is then assumed to be equal

to the load that causes a maximum stress in that panel equal to 2800 psi. The charts in the E06.51.13 draft are based on the glass failure model developed in [10]. In this model a glass panel is viewed as a weakest link system that will fail when the strength at any of

its points and in any direction is lower than corresponding applied normal stress. The strength

is assumed to be modeled by a two parameter Weibull distribution. The probability of failure of the panel is then calculated in accordance with Weibull 's classic theory. Conversely, it is possible to estimate from failure tests conducted on a sufficient number of panels the best fitting

parameters of the Weibull distribution of the strength. Comments on the possible effects of load capacity estimation methods and on differences between

values given in various documents are presented subsequently.

3.2.3 Effect of Statistical Errors Statistical errors are likely to have affected the values of the load capacities specified, e.g., by

the LOF charts. Note, in particular, that the methods of analysis used in [6] and in the E06.51.13 draft, and the respective charts, reflect

the influence of the aspect ratio of the plate on

the stresses and therefore on the failure loads, while the LOF charts [4], the old PPG charts [5], and the ASTM charts [7] do not. It is stated in

[4] that: "The ratio of the width and length of

a pane of glass ... may have an effect on its ability to withstand uniform loads. Test data show that adjustments for this ratio are small and follow no readily predictable pattern. . . . When the ratio of the long edge to the short edge does not exceed 5:1 any adjustment would not influence the choice of glass type or thickness." In the writers' opinion, the failure of [4] (as well as

[5] and [7]) to reflect the influence of the aspect

ratio is due to statistical errors. Such errors can be significant, since the coefficient of variation of the failure loads can be as high as 0.20 or so, and the number of specimens in any one

set of tests is, as a rule, relatively small (from

10 to 30, say). Statistical errors could thus easily obscure the aspect ratio effect. This point

will bo taken up again subsequently against the background of comparisons between graphs of various specified allowable loads. Statistical errors due to insufficient testing are also likely to affect values of the prestress used

in methods for estimating the load capacity of heat strengthened and tempered glass panels.

3.3 NBS Method for Estimating Load Capacity

A more recent method for estimating the load capacity of glass panels has been developed at the National Bureau of Standards (NBS). The method

58
entails two phases; obtaining basic strength

information, and estimating the load capacity of glass panels once this information is available.

The basic strength information consists of the parameters of the Weibnll distribution of the glass strength. It is possible to obtain this information from results of ring-on-ring tests performed on new glass, using a method based on fundamental fracture mechanics considerations,

described in [12]. The method for estimating the probability distribution of the load capacity of

the panel is also based on fundamental fracture mechanics considerations. Its essential features

are described in [13], and a listing of the computer program used to implement the method is included in [12] , The computer program used in the investigation presented in this work was an updated version of the program listed in [12] and is available from the authors.

An interesting result obtained in [12] is that the Weibull distribution of the glass strength, and the Weibnll distributions of the load capacity of the panels, have location parameters with positive

values, rather than with zero value, as assumed in [10]. This is an indication that the load capacities of panels have positive, rather than

zero, lower bounds. A similar result was obtained

by independent tests conducted at Nippon Sheet Glass Company in Ichihara City, Japan, and communicated to the writers by S. Kawabata. It was also shown in [12] that the procedure for estimating the load capacity of glass panels

developed at NBS is insensitive to the shape of flaws causing panel failure. Further research

conducted on the validity of the fracture mechanics

models used in this procedure is reported in [14],

in which these models were validated against

results of failure tests on about 100 annealed

glass panels with dimensions 1.52 m z 2.43 m z 6mm, conducted recently by the Ontario Research

Foundation [15]. Load capacities corresponding to the failure probability = 0.008 estimated for new glass by

the NBS method are shown in Figs. 4 to 8. For the heat strengthened glass the estimates were obtained by using strengths, based on ring-on-ring tests conducted on annealed glass specimens, augmented by a value assumed to be equal to the prestress due

to the heat strengthening process. The prestress

can be inferred from information on breaking

strengths tabulated for annealed and heat strengthened glass in [6]. The value of the prestress in the heat strengthened glass used in the calculations was 4,200 psi. A similar

procedure can be applied to tempered glass.

3.4 Comparisons Between Estimated Load Canacities

Figures 4 to 8 also include values of the load capacity corresponding to the failnre probability

Pp = 0.008 taken from the LOF charts [4] and from

the PPG charts [6]. Recall that the PPG values

from [6] pertain to glass assumed to have lost 33 percent of its mean strength owing to surface degradation in service. Approzimate values for the mean strength of new glass (based on the values

given for old glass in [6]) can be estimated as follows. In the case of annealed glass, the values

of [6] are multiplied by 1.5, i.e., by the ratio

(9000/6000) between the average breaking stress for new glass and for old glass assumed in the PPG calculations. This simple procedure assumes,

first, that stresses at the failure points are proportional to the applied load. Since it is indicated in [9] that the vast majority of annealed

glass panels were found to break in high stress regions, and since in such regions stresses are approz imate ly proportional to the loads, this assumption is acceptable to a first approzimat ion

.

A second assumption is that the coefficients of variation of the breaking stresses are the same for new and old glass. For heat strengthened glass the multiplying factor, rather than being 1.5, is assumed to be equal to the ratio between the sum

of the prestress and the breaking stress for new glass, and the sum of the prestress and the breaking stress for old glass. In both cases the values of the breaking stress corresponding to a failure probability = 0.008 are used. In Figs.

4 to 8 the PPG values from [6] are labeled 'PPG [6] (old glass) ', and their counterparts for new glass

are labeled 'PPG [6] (new glass)'. It is emphasized that these new glass strength values are only approzimate estimates.

It can be seen that the NBS estimates for new glass

are very close to the new glass values derived from

the PPG charts [6]. It is also noted that the influence of surface degradation on the estimated

allowable loads becomes smaller in the presence of prestressing due to heat treatment. Finally, note

that the general shapes of the curves ezpressing

the dependence of the load capacity upon aspect

ratio support the conjecture that, in the absence

of guidance by structural theory, this dependence

was clouded by sampling errors and was missed in LOF and other charts for this reason.

It is seen in Fig. 8 that there are instances where

the PPG values from [6] are higher than the LOF values, i.e., there does not appear to be a uniform pattern as far as the relative magnitudes of the values in the two sets of charts is concerned. An

ezamination of the LOF [4] and PPG [6] charts reveals that PPG values tend to be lower than LOF values (in some cases by as much as 50 percent) for the thinner annealed glass panels with relatively

small areas, and tend to be higher than LOF values

(in some instances by as much as 100 percent) for

the thicker annealed glass panels with larger areas

(i.e., about 80 ft^ or more), beat strengthened glass panels, and the thicker tempered glass panels

.

Comparisons between various estimates of the load capacities, such as those presented herein can reveal apparent anomalies that might be

attributable to deficiencies in the methods of analysis. For ezample, in spite of apparently

similar stress analyses and of the apparent use in developing the E06.51.13 draft of basic strengths lower than those implicit in [6], a 3/16 in. annealed glass panel with 10 ft^ area and aspect ratio 1.0 has a marginally higher load in the

59
E06.51.13 draft than in [6] (29 psf versns 28 psf ) , while for a 3/4 in. annealed glass panel with 100

ft^ area and aspect ratio 1.0 the E06.51.13 load

is considerably lower (42 psf versns 80 psf). Such

an inconsistency suggests that, in this case, at least one of the methods of analysis on which the two documents are based performs inadequately. One possible explanation is that the simplified model

on which [6] is based, which was briefly described earlier, may, for certain ranges of glass

thickness, panel area, and aspect ratio, result in load capacity estimates with unacceptable

deviations from the actual load capacities.

Another possibility is that, for certain ranges of these variables, unacceptable deviations maybe the result of forcing the location parameter of the Weibull distribution of glass strength to vanish,

as was done in [10] and in the E06.51.13 draft. This might bring the value of the variate

corresponding to the 0.008 percentage point closer

to zero than is actually the case. Further studies

of panel test results are necessary to verify the validity of such possible explanations.

4. RELIABILITY CONSIDERATIONS

4.1 Assessment of Past Practice

As stated earlier, the safety level inherent in a design criterion for glass cladding is regarded as acceptable if, in light of past experience with the behavior of glass facades, (1) it is satisfactory

from a public safety point of view, and (2) it does not entail wasteful overdesigns. A first and indispensable step in any attempt to develop new

design criteria is to assess past design practice

from these two points of view. Glass cladding started to develop as a major

architectural component primarily in the 1950's.

It would appear that for a period of at least 20

years a large number of designs have been carried

out on the basis of the criterion q^^ < q^^ # in which q^ denotes a conventional design wind load,

i.e., a design wind load taken from codes with wind loading requirements of the same type as those of

the Uniform Building Code (or codes such as the City of Atlanta Building Code) and q^^ denotes a conventional allowable load, i.e., an allowable

load taken from charts such as those issued by LOF

[4], PPG [5], or ASnU [7]. Such designs will be referred to here as conventional designs . Data on glass cladding failures due to high winds

are not readily available in a form suitable for analysis: however, it appears that, with the possible exception of a few local code

jurisdictions, failures have generally not occurred

owing to the inadequacy of provisions in building

codes and standards and glass design charts but,

rather, owing to such causes as: - design errors

- inadequate quality of panels being used

(due, e.g., to substandard manufacturing, as in the case of glass with nickel sulfide inclusions

causing spontaneous failure, or to mishandling of

the panels before or during construction)

- poor construction details

- the occurrence of exceptional storms such

as tornadoes the effects of which are deliberately

not accounted for in building codes

- wind—borne missiles, e.g., gravel blown by

wind from the roofs of adjacent buildings (this was apparently a major cause of damage to glass facades

in downtown Houston during hurricane Alicia). This suggests that, in general, conventional designs have been satisfactory from a public safety

standpoint. Should there be any reasons to doubt the acceptability of this assumption in any particular instance, a clear and well documented

case should be made for the opposite view.

As to the question whether conventional designs are acceptable from an economic standpoint, it can be argued that had they been viewed as wasteful, the probability of failure P^ = 0.008 associated with

the conventional allowable loads would have been

increased. To the writers' knowledge there have been no proposals to this effect.

4.2 Design Calibration: Case q^ = q^^

If it is assumed that conventional designs have been satisfactory from a safety standpoint, it follows that designs based on the inequality

q^ < q^, in which q^ = q^^ but q^^ is lower than

q^^ , would have safety levels higher than those inherent in conventional designs. This appears to be unnecessary, and would impose an undesirable

economic penalty on the glass cladding component

of building construction.

It has been argued that design values q^ < q should be specified for the following reason. If the failure probability corresponding to the conventional value q^^ is P^ = 0.008 after normal surface degradation (as understood in [5], see section 3.2 above), the failure probability will

be higher after the degradation that may occur

following 20 years of service. Thus, the use of

the conventional allowable load in design would not

ensure that the failure probability of the panel under the action of the design wind load will be

at most 0.008 during the entire anticipated life

of the panel. Thus, it has been argued that the allowable load used in design should be the value corresponding to a failure probability P^ = 0.008

of the panel degraded as apparently assumed in the E06.51.13 draft (i.e., with a loss of about 66 percent of the original strength).

In reality, 0.008 is not a 'magic number,' and it represents nothing more than a nominal value arrived at by experienced and competent

professionals within the context of conventional design practice . This value appears to have worked

well within that context, i.e., it has resulted in an acceptably small number of panel failures under wind loads. It is true that, implicit in

60
conventional practice are actnal values of the failnre probability under the load that are equal to or lower than 0.008 toward the beginning

of the panel’s service life, and possibly higher

than 0.008 at a later time. However, this does not matter as long as the corresponding number of panel

failures during the lifetime of the building is sufficiently low. Therefore, for design wind loads

q^ = , the use of allowable load charts based

on drastically reduced strengths (e.g., the charts

of the E06.51.13 draft) would not be justified.

3.3 Design Calibration; Case ^ q^^

part of the structure, that their dimensions are

4 ft X 6 ft, and that annealed glass is used. The

glass thicknesses corresponding in various charts

to the 45 psf load specified by the City of Atlanta

Building Code and to the 32 psf and 15.5 psf loads specified by ANSI A58. 1-1972 are listed below. % ^ ^wc ^w

" LOF [4] 1/4 in. 1974 PPG [5] 1/4 in. 1979 PPG [6] 5/16 in. E06.51.13 Draft 3/8 in.

3/16 in. 1/8 in. 3/16 in. 1/8 in. 1/4 in. 1/8 in. 5/16 in. 5/32 in.

It has been shown earlier that design wind loads calculated in accordance with the 1972 (or 1982) ANSI Standard A58.1 can be lower than conventional designs loads, e.g., for buildings in terrain

classified as urban, and/or for portions of the facades away from building edges. Design wind

loads incorporating aerodynamic and climatological

wind directionality effects also tend to be lower

than conventional wind loads, with the possible

exception of a few 'hot spots.’ First, we consider facades in which , by des ign , all

the panels have the same allowable load. For such

facades, the design criterion q^ < q^ becomes

max'fq^} < q^^. (2) where max'fq^} is the largest design wind pressure

on the facade (e.g., in Fig. 1, max'fq^} = 45 psf).

If max'fq^} > q^^., the design criterion given by Eq.

2 will result in a safe design. However, if maxfq^} < . the resulting design maybe unsafe,

i.e., the number of panel failures during the lifetime of the building would be higher than that inherent in a conventional design.

Second, we consider the case in which architectural

and construction considerations permit the division

of the facade into, say, two distinct zones with

panels having conventional allowable loads and

q^2c • largest design wind loads on the two zones are denoted by maxfq^j^} and max'(q^2 J* design inequalities are then used:

maxl(q^j) < q^jc (3a)

n'axfq^2^ < qa2c Usually max'fq^2

^ ^ ^wc * Therefore, even if maxlq^j] > q^^, , it still is possible that the total number of panel failures during the structure’s

lifetime will be higher than that inherent in a conventional design. This is of course even more strongly true if maxfq^j] < q^^.

.

Thus, unless the design wind load q^ > q^^ for all panels, the safety level inherent in a facade for which q^ = q^^ is lower than for a conventional

design. The safety level further decreases if

*la < <lac- For the sake of definiteness, we consider again

Figs. 1 and 2. We assume that the panels are designed on the basis of the loads on the upper

It can be seen that a conventional design would consist of annealed glass panels with 1/4 in. thickness. However, if the panels were designed

in accordance with the ANSI A58. 1-1972 Standard and the allowable loads were taken from [4] or [5], the panel thicknesses would be reduced to 1/8 in, over most of the facade and to 3/16 in. near the comers. Obviously, the expected number of panel failures during the lifetime of the building would

be considerably higher for this design than for the conventional design.

It therefore becomes understandable that following

the development of the ANSI Standard A58. 1-1972

some professionals may have felt uncomfortable with

situations such as the one just illustrated, and may have felt the need to counterbalance trends

toward less safe designs inherent, at least in some cases, in design provisions such as those of the ANSI A58.1 Standard. Even though the PPG Technical

Service Memo dated June 21, 1974 states that "PPG does not recommend ... safety factors," it is permissible to interpret PPG ’s changes in allowable

loads as changes in the factors of safety inherent

in glass cladding designs. The same could be said of the E06.51.13 draft. Assuming that the technical information included

in the revised PPG charts [6] or in the E06.51.13

draft was correct, the question arises whether the safety margins implicit in their respective choices

of glass strength (e.g., 67 percent of the value for new glass in the case of [6]) lead to acceptable safety levels for cladding glass

designs. For example, if the revised PPG charts

[6] are used in conjunction with the City of Atlanta Building Code, it can be seen from the glass thicknesses listed above the resulting design

would be uneconomical compared to a conventional

design. However, if those charts are used in conjunction with the ANSI A58. 1-1972 Standard, the resulting design would be less safe than that inherent in a conventional design. This is true because the thickness of the majority of the panels

(i.e., of all the panels not located in corner

zones) would be 1/8 in., instead of 1/4 in. as in

a conventional design, and the chances of failure

of some of those panels would therefore be greater

than for a conventional design.

It is seen that the disadvantage of, in effect, increasing safety factors by reducing the glass strength on which design charts are based is that such increases affect ^ indiscriminately designs

61
based on different loading criteria, so that while

the increases may be warranted for some loading criteria, they may well be unwarranted for others.

In the absence of guidance to the designer, and depending upon the loading criterion being used,

there would result in some instances facades that

are unnecessarily expensive; in some instances the facades would have acceptable safety levels; and

in some instances the resulting facades would be unsafe in relation to conventional designs.

4.4 Reliability-Based Glass Cladding Design

The existence of different criteria with respect

to design wind loads and to allowable panel loads inevitably creates confusion and dilemmas that have yet to be resolved. This can be done by an explicit use of modern structural reliability

procedures. Within the context of glass cladding

design, such a procedure has already been developed

and successfully applied (see, e.g., [3, Chapter

9] for a detailed description). For any given

glass cladding design of any given building, the procedure makes it possible to estimate the expected number of panel failures during the lifetime of the building. Conversely, given a specified safety level (i.e., a specified expected number of panel failures during the lifetime of the building), the procedure makes it possible to design the glass cladding so that it meets this safety requirement. The procedure is computerized,

and can account for the strength degradation of

the glass panels as a function of time. Its inputs consist of relevant wind climate and aerodynamic

information, and of the mean value and the coefficient of variation of the load capacity of

the glass panels

.

By applying the reliability-based procedure just mentioned to a sufficient number of typical

designs, it is possible to obtain information on

the expected number of panel failures per lifetime

inherent in those designs. Such information can

then be used to establish safety level criteria (i.e., acceptable expected numbers of panel

failures per lifetime). The latter would serve as

a basis for risk-consistent design criteria for glass cladding subjected to wind loads, which would supersede the present unsatisfactory state of affairs. Specifications involving the use of 'design factors' of the type included in the LOF

[4] or PPG [5] charts could be contemplated. A research effort on practical methods for achieving risk-consistent designs, involving the National

Bureau of Standards, Colorado State University, and Gordon B. Smith Corporation, has recently been

planned

.

4 . 5 Comments on a Future ASTO Standard

The evolution of consistent criteria outlined above

can take time, and should be gradual. The effort currently being undertaken by the ASTM E06.51.13

task group, whose scope is limited to preparing a standard concerning the load capacity of glass panels, could contribute usefully to such an evolution. For this to be the case, such a standard should meet the following requirements:

- Its charts should be derived from a basic strength of glass (as expressed, e.g., by the parameters of the probability distribution of the strength of glass) (a) defined explicitly in terms

of the basic strength of glass implicit in the ASTM

E 330 Standard [7] or of the basic strength of new

glass, and (b) meeting with the explicit approval

of the ASTM E06. 51.13 task group. - It should be based on an acceptable method

of calculation. In the writers' opinion, the method developed in [10] and used in the E06.51.13 draft

would be acceptable if it were modified to be compatible with experimental evidence that the lower bound of the glass strength is not zero, as assumed in [10] (i.e., that the assumed Weibull distribution of the glass strength has three, rather than two parameters). As mentioned earlier,

the assumption that the glass strength is described probabilistically by a two-parameter Weibull distribution could have significant effects on the extrapolation of test results to low probabilities

of failure (e.g., = 0.008).

- Allowable loads included in the standard, calculated by the method of [10] or by any other acceptable method of calculation, should be verified for an acceptable number of types of panels against reliable test results.

- The information on load capacity included

in the standard could, if desired, consist of the 0.008 percentage point. However, in addition to the 0.008 percentage point, the standard should

include the mean values and the coefficients of variation of the load capacities as well. Failure

to include such information would seriously impair the usefulness of the standard in achieving

rational and risk-consistent designs, and would represent a step backward with respect to the LOF

and PPG charts, which include useful probabilistic

information in addition to the 0.008 percentage

point.

5. SUMMARY AND CONCLUSIONS The safety and cost of glass cladding facades subjected to wind loads depends upon the criteria

used for their design. These criteria specify the design wind loads and the allowable panel loads.

In this paper a review has been presented of: the evolution in recent years of methods for the estimation and specification of design wind loads for building facades; recent developments

concerning the estimation and specification of allowable loads on glass cladding panels, and controversies surrounding those developments; and recent developments in structural reliability as applied to glass cladding, which allow rational decisions to be made concerning the adoption of satisfactory design criteria specifying both design

loads and allowable panel loads. In view of recent changes in methods for estimating design wind loads and allowable panel loads, it would be imprudent or wasteful to adopt a standard specifying allowable panel loads without giving due consideration to the overall reliability aspects

of the cladding glass design problem.

62
6. ACKNOWLEDGMENT

The writers wish to acknowledge the capable computational assistance of William A. Nolan of the Department of Civil Engineering, The Johns Bopkins

University.

7. REFEBENCES

(1 ) American National Standard A58.1, Bnilding Code Reonirements for Minimum Design Loads in Buildings

and Other Structures . National Standards Institute, New York, N.Y., 1972.

(2) Ravindra, M.K., Cornell, C.A., and Galambos,

T.V., 'Wind and Snow Load Factors for Use in LPFD, ' Jonrnal of the Structural Division . ASCE, Vol. 104,

1978, 1443-1457.

(3) Simiu, E., and Scanlan, R.F., Wind Effects on

Structures . Wi ley-Interscience , New York, N.Y.

,

1986.

(4) LOP Technical Information — Strength of Glass

Under Wind Loads . 805-109, Toledo, Ohio, January,

1980.

(5) PPG Industries Glass Product Recommendations;

Wind Load Performance . Technical Service Report No. lOlA, PPG Industries. Pittsburgh, 1974.

(6) PPG Glass Thickness Recommendations to Meet Architects Specified 1-Minute Load . Technical Services /Flat Glass Division, PPG Industries,

Pittsburgh, April 23, 1979.

(7) ANSI/ASTM E330 - 70. Standard Test Method for Structural Performance of Exterior Windows Curtain

Walls, and Doors Under the Influence of Wind Loads . 1978 Annual Book of ASTH Standards, American

Society for Testing and Materials, Philadelphia,

1978.

(8) Glass and Uniform Loads . LOF Technical

Information, ATS-108, File #1 - Strength of Glass,

Toledo, Ohio, April 30, 1979.

(9) Krall, W.R., et.al.. 'The Behavior of Float

Glass Under Uniform Wind Loading,' Proceedings , Fourth U.S . National Conference on Wind Engineering

Research, Department of Civil Engineering.

University of Washington, Seattle, 1981.

(10) Reason, W.L. , and Morgan, J.R. ,

i 'Glass Failure Prediction Model,' Journal of Structural

Engineering . ASCE, Vol. 110, 1984, 197-212.

(11) Somes, N.F., Dikkers, R .D . , and Boone, T.H., Lubbock Tornado; A Survey of Damage in an Urban Area . Technical Note 558, National Bureau of Standards, Washington, D .C , , March 1971.

(12) Simiu, E., et al., Ring-on-Ring Tests and Load Capacity of Cladding Glass . NBS Building

Science Series 162, National Bureau of Standards, Gaithersburg, Md . , August 1984.

(13) Reed, D .A . , and Simiu, E., 'Wind Loading and Strength of Cladding Glass,' Journal of Structural Engineering . ASCE, Vol. 110, 1984, 715-729.

(14) Simiu, E., Nolan, W .A . , and Lechner, J.,

'Glass Fracture Phenomenology and Loading Rate,' submitted for publication to the Journal of Structural Engineering . April. 1986.

(15) Johar, S., Dynamic Fracture of Glass. Part III . Ontario Research Foundation, Mississauga,

Ontario, Canada, July. 1982.

63
TOP FLOOR

DESIGN PRESSURE DISTRIBUTION REO^JIRED CITY OF ATLANTA BUILDING CODE

(All Loads in PSF) SIDE

leeward

SIDE

DESIGN PRESSURE DISTRIBUTION TO BE SPECIFIED FOR

WINDOWS ON ALL ELEVATIONS (ALL LOADS IN PSF)

LEEwaRD «vihO*VARO

Figure 1. City of Atlanta Building

Code wind loads (from PPG Technical Service Memorandum, June 21, 1974).

Figure 2. ANSI A58. 1-1972 wind loads (from PPG Technical Service Memorandum

June 21. 1974).

64
Figure 3. Dependence on direction of the pressure coefficient near a building corner.

65
1

“ a

S‘

8

- o

8

-

- 28'

8

-

ANNEALED GLASS PANELS

SIMPLY SUPPORTED ON FOUR SIDES AREA - 15 SQUARE FEET

T - 1/4 INCHES

o

A+

NBS PPG (6) (old glass) LOF {41 PPG (6) (new glass)

Figure 5
8

-

8

-

i-o -

- t. . d 8- lh\

S- £

“ b«?J •o 8- ^ i;

R-e

8

-

e

IIHEAT STRENGTHENED GLASS PANELS

SIMPLY SUPPORTED ON FOUR SIDES

AREA - 15 SQUARE FEET

T - 1/4 INCHES

NBS o PPG (61 (old glass) A LOF (4) + PPG [61 (new glass)

Figure 7

67

THEME II

Earthquake Engineering

SPECTRAL CHARACTERISTICS OF GROUND MOTION

IN THE TOKYO METROPOLITAN AREA

BY Keiichi Ohtani , and Shigeo Kinoshita

National Research Center for Disaster Prevention

Science and Technology Agency

ABSTRACT

The authors propose a tentative estimation

procedure to estimate the average amplification

characteristics of a sedimentary layer with a thickness of several kilometers for the short period S-waves (0.2-2 s) in a specific region

of the Tokyo metropolitan area. The average amplification characteristics are given by the average of ratios of the zero-damped velocity

response spectra of surface strong-motion to those of bedrock motion. The zero-damped

velocity response spectra of bedrock motion are given by the models which were constructed by

using the data of accelerometers installed at

the bottoms of deep boreholes in the pre- Tertiary basement. The results obtained by

using our method are compared with the calculated wave- transfer function of the thick

sedimentary layer for the SH-waves. INTRODUCTION

To investigate the earthquake response of the alluvium, observation of the earthquake motions

by using seismometers installed on ground and

under the alluviiom by a borehole, was performed. The thickness of the alluvium is a hundred meters; in the Tokyo metropolitan area, borehole observation was practical and economical. During the last ten years, the earthquake

response of the alluvium was investigated using

borehole observation at many sites in this

area. Recently, in Japan, we learned that the pre- Tertiary basement is seismic bedrock. As a result, it is necessary to investigate the earthquake response of thick sedimentary layers

of several kilometers of bedrock. The method

used to investigate the earthquake response of

the alluvium became impractical because

observations of earthquake motions in the pre- Tertiary basement by means of boreholes are expensive and difficult to perform. Instead of using bedrock motion data, which are directly observable by deep boreholes, it is economical and acceptable to use a statistical

model to estimate the spectral characteristics

of bedrock motions. In this report, as a tentative method to estimate the earthquake

response of thick sedimentary layer, we examined a method that used the data of the surface strong-motions and models for

estimating the spectral characteristics of bedrock motions. The models are constructed

using acceleration records obtained at the bottom of three deep boreholes in the Tokyo

metropolitan area.

2. METHOD

The amplification characteristics of the sedimentary layer are defined as G(T) - SVq(T) / SVi(T), where SV£(T) and SVq(T) are the zero-damped

velocity response spectra at the period T of

the bedrock motion and the surface ground

motion, respectively. SVj^(T) is obtained from

the practical model and SVq(T) is obtained from

the data observed on the surface ground. The average amplification characteristics of G(T)

is estimated by the following relation;

- I

SVo(T; M^, R^) / SV^ (T; M^, R^) ,

where M^ and R^ are the JMA magnitude and the hypocentral distance (in km^ of the n-th

earthquake, respectively. G(T) means the average of earthquake response of the thick sedimentary layer at the specific site. The zero-damped velocity response spectrum SV^(T;

M, R) of the bedrock motions is estimated by

the following relations;

SVi(T; M, R) - ioa(T)M-b(T)log3^o^+c(T)

The coefficients a(T) , b(T) and c(T) at each period T, are given in Fig. 1. Model I is obtained by using the earthquakes whose epicenters were located out of the volcanic

front in Japan and Model II is obtained by the earthquakes whose epicenters were located in and around the volcanic front. The results in

Fig. 1 were constructed by using the acceleration data obtained at the bottom of deep boreholes in the three sites shown in Fig. 2; Iwatsuki (IWT, GL-3510 m) , Shimohsa (SHM, GL- 2300 m) and Fuchu (FCH, GL-2750 m)

.

3. RESULTS

The average amplification characteristics of sedimentary layer at the sites were investigated; Iwatsuki (IWT; 35“ 55' 33"N, 139° 4A' 17"E), Urayasu (URAYA SU; 35° 39' 00"N, 139°

71
54' 19"E) and Tama (NT2; 35° 37' 65"N, 139° 26' 09 "E) . The earthquake data are shown in Table

1. These earthquakes belong to the earthquakes

used to construct the Model I of bedrock

motions. Table 2 shows the estimated velocity

structure at three sites. Estimated average amplification characteristics of sedimentary

layer are shown in Fig. 3 (IWT), Fig. 4 (URAYASU) and Fig. 5 (NT2) . G is the estimated

average amplification characteristics. is the gain of transfer function of sedimentary

layer calculated by using the velocity structure in Table 2. SGq is the gain of transfer function of surface alluvium.

The results show that the average amplification

characteristics are mainly influenced by the response of sedimentary layer for the SH-waves

and have maximum amplification factors of several ten times. Also, the average amplifi- cation characteristics have an envelope similar

to the amplification characteristics of soft alluvium for the SH-waves.

4. CONCLUDING REMARKS

Based on the models used to estimate the spectral characteristics of the bedrock

motions, we proposed a tentative method to estimate the average amplification characteristics of sedimentary layer for the short period

S- waves. The results obtained indicate that

our tentative method may be used to estimate

the average response characteristics of sedimentary layer in the Tokyo metropolitan

area.

72
Table 1-1 Data of earthquakes for IV'n? site

No. Date

1 Oct. 5, 1977

2 Dec

.

17, 1977

3 Feb. 20, 1978

4 Mar. 20, 1978

5 Jul. 11, 1979

6 Sep. 24, 1980

7 Sep. 25, 1980

8 Jul. 23, 1982

9 Feb. 27, 1983

Lati tude Longitilde 36 '’08' 139 '’52' 36'’35' 141-’05' 38M5' 142“12'

36 ’05’ 139°53‘

36“36* 141019 . 35'“58’ 139“48’

35°31’ 140 '’13' 36»11’ 141‘’57’ 35“54 ' 140°09*

Depth

(km) M^g^^tude

60 5.5

50 5.6

50 6.7

60 5.5

40 5.9

80 5.4

80 6.1

30 7.0

70 6.0

Table 1-2 Data of earthquakes for URAYASU site

No. Date Latitude Longitude Depth

(km) Magnitude

1 Sep. 24, 1980 35“58' 139“48‘ 80 5.4

2 Sep. 25, 1980 35 ‘’31’ 140 ‘’13’ 80 6.1

3 Jul. 23, 1982 36<*11' 141»57’ 30 7.0

4 Feb. 27, 1983 35'>54’ 140 °09' 70 6.0

5 Jan.> 17, 1984 36°27’ 141“ 15' 40 5.6

6 Jan., 18, 1984 36“27' 141'’16’ 40 5.9

7 Sep. 19, 1984 34°03' 141“33’ 10 6.6

Table 1-3 Data of earthquakes for NT2 site

No. Date Latitude Longitude Depth

(km) Magnitude

1 Mar. 20, 1978 36 ’05' 139°53' 60 5.5

2 Jun

.

12, 1978 38°09’ 142O10' 40 7.4

3 Sep., 24, 1980 35“58' 139°48' 80 5.4

4 Sep. 25, 1980 35“31’ 140°13' 80 6.1

5 Feb. 27, 1983 35“54' 140 <’09' 70 6.0

6 Jul. 2, 1983 36°54 141°12’ 50 5.8

7 Sep.• 19, 1984 34“03' 141°33' 10 6.6

73
Table 2-1 Estimated velocity structure at IWT site Layer No. S-wave velocity

(km/s) Density

(g/cm^) Thickness

(m) 1 0.189 1.7 5.0

2 0.155 1.7 6.0

3 0.3 1.8 6.0

4 0.18 1.7 2.0

5 0,54 1.9 181.0

6 0.78 2.0 750.0

7 1.17 2.2 800.0

8 2.5 2.5 Table 2-2 Estimated velocity structure at URAYASU site Layer No

.

S-wave velocity

(km/s) Density

(g/cm^) Thic)cness

(m) 1 0.14 1.7 3.0

2 0.2 1.8 5.0

3 0.12 1.7 15.0

4 0.15 1.7 9 ,0

5 0.2 1.8 4.0

6 0.45 1.9 264.0

7 0.67 2.0 1000.0

8 1.5 2.2 1000.0

9 2.5 2.5 Table 2 -3 Estimated velocity structure at NT2 site Layer No. S-wave velocity

(km/s) Density

(g/cm^) Thi ckness

(m)

1 0,16 1.7 12.5

2 0.13 1.8 2.5

3 0.33 1.9 27.0

4 0.37 1.8 9.0

5 0.33 1.8 14.0

6 0.41 2.0 17.0

7 0.36 1.9 20.0

8 0.44 2.0 218.0

9 0.76 2.1 670.0

10 1.3 2.5 980.0

11 1.6 2.6 830.0

12 2.5 2.7

74
al

log iy

139*E 141‘E

Fig. 2 Location of deep borehole observatories

75
Gain

Gain

It

Estimated average amplifi- cation characteristics of sedimentary layer (IWT)

Estimated average amplification characteristics of sedimentary layer (URAYASU) f

Period (s)

76
Fig. 5 Estimated average amplification characteristics of sedimentary layer (NT2)

77
STRONG-MOTION RECORDINGS FROM THE

HOLLISTER, CALIFORNIA, AREA M5.5 EARTHQUAKE,

26 JANUARY 1986

BY

A. Gerald Brady, Richard P. Maley,

and G. Noel Bycroft^

ABSTRACT On January 26, 1986, a moderate earthquake

(magnitude Berkeley; Caltech) occurred approximately 12 km SE of Hollister,

California. The coordinates of the earthquake's epicenter were 36° 48.32' North

and 121° 16.94' West, which is located along

the southern end of the little-known Quien

Sabe fault, 7 km east of the Calaveras Fault. The right-lateral strike slip event occurred

at a depth of 8.2 km. Among the stations operated by the United

States Geological Survey (USGS) that recorded

this event are:

1. Five stations in the Hollister vicinity, with film recorders.

2. The Hollister Digital Differential

Array, located at the airport 4 km north

of Hollister.

3. Six stations of the Bear Valley array, spreading over an area reaching 46 km

south of Hollister, with film recorders. The main purpose of this report is to bring to the attention of the earthquake engineering

and scientific communities the existence of a set of film and digital recordings of the earthquake from strong motion accelerographs

operated and maintained by the USGS. A preliminary analysis is made of the digital

recordings and their relationships.

KEYWORDS: accelerograms; earthquake; processing.

1.

INTRODUCTION

On January 26, 1986, 1920:50.9 GMT, a moderate earthquake (magnitude Ml 5.5, Berkeley; Caltech; Mb5.2, NEIS) occurred approximately 12 km SE of Hollister, California.

The coordinates of the earthquake's epicenter

were 36° 48.32' North and 121° 16.94' West, which is located along the southern end of the Quien Sabe fault of D. Herd (unpubl.), 7 km

east of the Calaveras Fault. The event occurred at a depth of 8.2 km and its movement

was right-lateral strike slip. (Rob Cockerham, USGS, oral communication, 1986). Among the stations operated by the United

States Geological Survey (USGS) (Ref. 1) that

recorded this event are (see Table 1):

1. Stations near Hollister (see Fig. 1).

2. the Hollister Digital Differential

Array, located at the airport 4 km north

of Hollister (Ref. 2 and Fig. 2).

3. the Bear Valley array, spreading over an area reaching 46 km south of Hollister

(Ref. 3). The main purpose of this report is to bring to the attention of the earthquake engineering

and scientific communities the existence of a set of film and digital recordings of the earthquake from strong motion accelerographs

operated and maintained by the USGS. Although

the report is not complete, the earthquake

occurred sufficiently close to the Western

Region Headquarters of the USGS in Menlo Park

that most of the triggered records were recovered within a few days, and some preliminary analysis was possible on the digital recordings. Updated information on the earthquake,

records, and processing will be provided in subsequent data reports. Further study of the

set of records from the Bear Valley array will

be required before a decision is made on their digitization. As for all earthquakes (for example, San Fernando, 1971) the more remote triggered stations are of much less interest,

and their digitization and processing is less cost-efficient than those with accelerations

of 10% g or higher. This report contains copies of the film records obtained from 11 of these stations and processed records from the Hollister array. The stations at Anderson Dam and Pinnacles did

not trigger. The San Justo Dam instruments

are maintained by the USGS under a cooperative

agreement with the Bureau of Reclamation.

The Hollister array (located 11 km from the San Andreas Fault) is a 24-channel, two-dimensional system with two legs at 33° separation

with accelerometers placed at 0, 61, 457, and 610 m on one leg, and 244 and 305 m on the second. The sensors are Kinemetrics triaxial

force balance accelerometers mounted at the center of a 2.1 m diameter concrete pad, well coupled to the soil. The system has a 2.5 second pre-event memory (allowing the record-

^U.S. Geological Survey, Menlo Park, California.

78
ing of pre-trigger ing onset motions) and is triggered by vertical motions exceeding 0.01 g in the range of 1 to 10 Hz. The data are digitized at a rate of 200 samples per second

with a total recording time capability of 20 minutes. Two additional triaxial accelerographs, including the Differential Array film

recorder, are mounted on an isolated pier in

the recorder house.

2. DESCRIPTION OF RECORDS Epicentral and hypocentral distances for the stations, rounded to the nearest kilometer,

are given in Table 1. Table 2 lists the scaled peak accelerations for each component. Film records from the Hollister area have epicentral distances ranging from 11 to 16 km, with peak horizontal acceleration of 0.17 g (Damler Residence) and vertical acceleration

of 0.29 g (Hollister City Hall Annex). Copies

of the first 20-25 seconds of these film records are shown in Figure 3. Half-second

time marks appear at the lower edge. Digital records from the Hollister Differential Digital Array, 15 km from the epicenter,

have peak horizontal accelerations ranging

from 0.09 to 0.12 g, and vertical, 0.08 to 0.17 g. Station 2 failed to operate correctly

(A/D converter problem) and one horizontal

channel of Station 4 failed. Film records from the Bear Valley Array, with epicentral distances ranging from 17 to 37 km, exhibit peak horizontal accelerations ranging

from 0.12 g down to the approximate triggering

level at 0.01 g. The Pinnacles and Anderson Dam stations did

not trigger. Further distant stations have

not been visited for this report.

3. DIGITIZATION AND PROCESSING

The digital records from the Differential

Array are processed by the following steps. Further details may be found in Ref. 4.

1. The uncorrected data are prepared by subtracting the mean. The instrument

sensitivities, determined by peak-to-peak

voltages, are used to scale the ordinates to acceleration.

2. The data are passed through a correction

algorithm that applies a 50 Hz (high-frequency) low-pass filter, instrument correction, base-line correction in the form of a 0.2 Hz low-frequency (5 second, long-period) filter,

and decimation to 200 sps. This initial preliminary selection of 5 seconds for Che

long-period filter is based on the convention

of retaining a period content somewhat longer

than the strong-motion duration of the records

(Ref. 5), here estimated at 2-3 seconds.

Fig. 4 shows the preliminary calculated

displacements from the Differential Array

stations, arranged by component, and aligned

for simultaneous timing. Further analysis and processing is required to differentiate

between the differences that are due to noise

and the differences that the array is designed

to study.

4. SUMMARY Although in this preliminary report Che quoted values for the earthquake epicenter may even- tually be altered, it appears that a significant set of records has been obtained from Che magnitude 5.5 event. Current epicentral

distances indicate five records to the west within 15 km. A set of five digital records obtained at the Hollister array are 15 km from

the epicenter. A further set of film records

to the southeast has been recovered at epicentral distances in Che range 17-37 km. Digitization and further processing of the film records will await a decision on the overall significance of the event. Further

corrections applied to Che digital recordings,

and investigations on the total ground motion

at the Hollister array will be described in subsequent reports.

5. ACKNOWLEDGEMENTS

We appreciate the kindness of the property

owners, where Che Bear Valley Array instru- ments are located, for permission originally

to install the array and subsequently annually

to maintain it. The cooperative agreement

with the Bureau of Reclamation has resulted in the San Justo Dam records, from instruments

installed prior to the dam being built.

6. ATTRIBUTIONS

A. G. Brady leads the Data Management project with P. N. Mork and B. L. Silverstein; R. P. Maley heads the Network Operations project with E. C. Etheredge and technicians D. A. Johnson, A. V. Acosta, R. D. Forshee and M. J. Salsman. Text, figures. Cables, and records were prepared through the efforts of all.

7.

APPENDIX

The appendix contains three tables listing processed records since 1978.

79
8. REFERENCES

(1) Switzer, L. , Johnson, D., Malay, R. , and Matthiesen, R. , 1981; Western Hemisphere

Strong-Motion Station List - 1980; U,S. Geological Open-File Report No.- 81-664, 162 p. (2) Malay, R. P., and Etheredge, E. C., 1984; The Development of Ground and Structural Response Strong-Motion Instrumentation Arrays

in the United States: Eighth World conference

on Earthquake Engineering, San Francisco, California, 8 p. (3) Dielman, R. J. , Hanks, T. C., and

Trifunac, M. D., 1975; An array of strong-motion accelerographs in Bear Valley, California; BSSA, v. 65 no. 1. (4) Converse, A. M. , 1984; AGRAM: A Series o Computer Programs for Processing Digitized

Strong-Motion Accelerograms, Version 2.0: U.S. Geological Survey Open-File Report

81-525, 118 p. (5) Basili, M, , and Brady, A. G., 1978; Low

Frequency Filtering and the Selection of Limits for Accelerogram Corrections: Sixth European Conference on Earthquake Engineering

Dubrovnik, Yugoslavia.

Table 1.- Epicentral and Hypocentral Distances Earthquake coordinates: 36.805°N, 121.282°W

depth: 8.2 kilometers

Station Name Latitude

(North) Longitude(West) EpicentralDistance(km) HypocentralDistance(km) Azimuth from Epicenter

Bear Valley 5 Callens Ranch36.673 121.195 17 19 152.10 Bear Valley 1 CDF Fire Station

36,573 121.184 27 28 161.22 Bear Valley 6 James Ranch

36.504 121,101 37 38 154.18 Bear Valley 7 Pinnacles (a) 36.483 121.184 37 38 165.68 Bear Valley 2 Stone Canyon West

36.636 121.234 19 21 167.10 Bear Valley 14 (b) Upper Butts Ranch

36.569 121.043 34 35 140.85 Bear Valley 10 Webb Residence

36.532 121.143 33 34 157.72 Bear Valley 12 Williams Ranch

36.658 121.249 17 19 169.71 Anderson Dam (a) 37.166 121.626 50 51 322.83 Hollister

City Hall Annex

36,851 121.402 12 14 295.53 Hollister

SAGO Vault

36.765 121.446 16 18 251.40 Hollister Damler Residence

36.807 121.408 11 14 272.65 Hollister

Differential Array

36.888 121.413 15 17 308.37 San Justo Damsite Abutment

36.815 121.447 15 17 274.37 San Justo Damsite

Dike 36.827 121.445 15 17 279.45

Notes: (a) No trigger

(b) Not yet visited
Table 2.- Peak Accelerations, scaled from original film or preliminary

digital processing

Station Station Name Number Components Scaled peaks (g)

L V T L V T Bear Valley Array: Callens Ranch 1474 310 Up 220 0.05 0.06 0.04 CDF Fire Station 1210 310 Up 220 0.04 0.02 0.04 James Ranch 1475 310 Up 220 0.02 0.01 0.01 Pinnacles (a) 1476 310 Up 220 Stone Canyon West 1343 130 Up 040 0.03 0.02 0.04 Upper Butts Ranch (b) 1483 310 Up 220 Webb Residence 1479 310 Up 220 0.04 0.01 0.04 Williams Ranch 1481 310 Up 220 0.12 0.06 0.12 Hollister

City Hall Annex 1575 180 Up 090 0.10 0.29 0.12 SAGO Vault 1032 360 Up 270 0.04 0.04 0.04 Damler Residence 1657 360 Up 270 0.17 0.09 0.14 Hollister Differential Array 1656 Station 1 255 Up 345 0.09 0.17 0.11 Station 2 255 Up 345 — — —

Station 3 255 Up 345 0.10 0.12 0.10 Station 4 255 Up 345 0.10 0.08 —

Station 5 255 Up 345 0.11 0.17 0.12 Station 6 255 Up 345 0.10 0.15 0.11 Recorder building 255 Up 165 0.10 0.15 0.10 San Justo Damsite 1655 Abutment 360 Up 2 70 0.16 0.07 0.14

Dike 360 Up 270 0.09 0.04 0.08

Notes

:

(a) No trigger

(b) Not yet visited

(c) No operation

(d) No op, one channel

81
and adjacent faults.

82
83
HOLLISTER: CITY HALL ANNEX

180

HOLLISTER DIFFERENTIAL ARRAY

HOLLISTER: DAMLER RESIDENCE ^

BUREAU OF RECLAMATION

SAN JUSTO DAMSITE: DIKE

360

I < SAN JUSTO DAMSITE: ABUTMENT

360

Figure 3. Copies of the first 20-25 seconds of the film records. Most ran for at least 60 seconds for the WWVB time record. Refer to Taole 1 for peak values

84
iCCSMOS

seconos Figure 4. Hollister array aligned displacements from preliminary accelerogram

correction, using band-pass filter from 50 Hz to 5 sec.

85
APPENDIX

CURRENT LIST OF PROCESSED RECORDS January, 1987 uses processing of records from the USGS permanent network of strong-motion

accelerographs and associated networks. Strong motion data from earthquakes 1978 and later including an isolated earlier event. TABLE Al. Chronological list of events and associated reports describing the existence/processing/analysis/availability of digital data on tape, or

at the National Strong Motion Data Center in Menlo Park.

Date & Time Earthquake Reference (see attached list)

January 1, 1975; 0355 G>fr Southern Alaska OFR 86-191 (Ref. 16) March 25, 1978 Coyote Dam, California

OFR 83-166 (Ref. 6)

August 27, 1978 and two later shocks Monticello Dam, Jenkinsville , South CarolinaOFR 81-0448 (Ref. 2)

August 6, 1979 Coyote Lake, CaliforniaOFR 81-42 (Ref. 4)

October 15, 1979 Imperial Valley, CaliforniaOFR 80-703 (Ref. 5)

October 15, 1979 Imperial Valley, California OFR 82-183 (Ref. 11)

October 15, 1979; 2317:41,

2318:20, 2318:40 GMT Imperial Valley California aftershock

OFR 86-441 (Ref. 3)

October 16, 1979, 0706 GMT Monticello Dam, Jenkinsville,

South Carolina

OFR 81-1241 (Ref. 10)

December 13, 1981 and March 18, 1983 Solomon Islands OFR 86-264 (Ref. 15)

February 13, 14, and 23, 1983 Monasavu Dam, Fiji OFR 85-375 (Ref. 12)

May 2 and May 9, 1983 Coalinga, California OFR 84-626 (Ref. 9) July 9, 1983; 0740 GMT Coalinga, California OFR 85-584 (Ref. 13) July 22, 1983; 0239 GMT Coalinga, California OFR 85-250 (Ref. 14) April 24, 1984 Morgan Hill, California

OFR 84-498, Vol I and II (Refs. 7 and 8) (Compiled by Seena Hoose) December 23, 1985; 0516 GMT

and Nov. 9, Dec. 23, Dec 25

Northwest Territories,

Canada

OFR 86- , (Ref. 17)

January 26, 1986; 1920 GMT Hollister, California

OFR 86-156, (Ref. 1)

86
TABLE A2. Processed records in each report

January 1, 1975; 0355 GMT; southern Alaska; OFR 86-191. Records (4): Anchorage, 500 W. Third St., Basement

Anchorage, Alaskan Methodist University

Anchorage, Government Hospital

Talkeetna, FAA-VOR Building March 25, 1978; Records (3); Coyote Dam, California; OFR 83-166. Coyote Dam, Ukiah, California: abutment, toe, crest August 27, 1978, 1023 GMT and 2 later shocks; Monticello Dam, South Carolina,

OFR 81-0448. Records (3): Jenkinsville , Monticello Dam

Shared abutment (center crest) August 6, 1979, Coyote Lake, California; OFR 81-42. Records (6): Coyote Creek, San Martin, California

Gilroy Array: Station 6, San Ysidro

Gilroy Array: Station 4, San Ysidro School Gilroy Array: Station 3, Sewage Treatment Plant Gilroy Array: Station 2, Mission Trails Motel Gilroy Array: Station 1, Gavilan College

October 15, 1979, 2317 GMT; The Imperial Valley Earthquake; OFR 80-703. Records (22): El Centro Array 7, Imperial Valley College

El Centro Array 6, Huston Road

El Centro, Bonds Corner, Hiways 98 & 115 El Centro Array 8, Cruickshank Road

El Centro Array 5, James Road

El Centro Differential Array

El Centro Array 4, Anderson Road

Brawley, Brawley Municipal Airport

Holtville, California, Holtville Post Office

El Centro Array 10, Keystone Road

Calexico, California, Calexico Fire Station

El Centro Array 11, McCabe School

El Centro Array 3, Pine Union School Parachute Test Facility

El Centro Array 2, Keystone Road

El Centro Array 12, Brockman Road

Calipatria, California, Calipatria Fire Station

El Centro Array 13, Strobel Residence

El Centro Array 1, Borchard Ranch Superstition Mountain, California

Plaster City, California, Storehouse Coachella Canal Number 4, California

October 15, 1979, Records (6)

October 15, 1979, Records (6)

2317:41 GMT; Imperial Valley Aftershocks; OFR 86-441

El Centro Array 5, James Road

El Centro Array 6, Huston Road

El Centro Array 7, Imperial Valley College

El Centro Array 8, Cruickshank Road

El Centro Array 9, Commercial Ave.

El Centro Differential Array

2318:20 GMT; Imperial Valley Aftershocks; OFR 86-441

El Centro Array 5, James Road

El Centro Array 6, Huston Road

El Centro Array 7, Imperial Valley College

El Centro Array 8, Cruickshank Road

El Centro Array 9, Commercial Ave.

El Centro Differential Array

87
TABLE A2. Processed records in each report (continued)

October 15, Records

October 15, Records

October 16, Records

December 13 Records

February 13 Records

1979, 2318:40 GMT; Imperial Valley Aftershock; OFR 86-441

(7) El Centro Array 6, Huston Road

El Centro Array 7, Imperial Valley College

El Centro Array 8, Cruickshank Road

El Centro Array 9, Commercial Ave. El Centro Differential Array

Bonds Corner, Highways 115 & 98 Holtville Post Office

1979; The Imperial Valley, California; OFR 82-183;

(22): This report contains the time-dependent response spectrum

plots for the same records in OFR 80-703, above.

1979, 0706 GMT, Monticello Dam, South Carolina, OFR 81-1214.

(1): Jenkinsville, South Carolina, Monticello Dam

shared abutment (center crest)

1981 and March 18, 1983; Solomon Islands, OFR 86-264

(5): Dec. 13, 1981, 0129 GMT: 460 Beach, Panguna Mine, Bougainville Island.

Dec. 13, 1981, 1324 GMT: March 18, 1983: Arawa Town

Bato Bridge

BVE80, Panguna Mine.

14, and 23, 1983; Monasavu Dam, Fiji; OFR 85-375

(3): Feb 13, 14, 23, 1983: Monasavu Dam. May 2 and May 9, 1983; Coalinga, California; OFR 84-625. Records (13): May 2, 1983, 2342 GMT: Pleasant Valley Pump Plant: switchyard, basement

May 9, 1983, 0249 GMT Anticline Ridge: freefield and pad Burnett Construction

Oil City

Oil Fields Fire Station

Palmer Avenue

Skunk Hollow

Pleasant Valley Pump Plant: switchyard, basement,

1st floor, roof July 9, 1983; 0740 GMT; Coalinga, California; OFR 85-584 Records (9): Anticline Ridge: freefield and pad Burnett Construction

Oil City

Oil Fields Fire Station: freefield and pad Palmer Avenue

Skunk Hollow

Transmitter Hill July 22, 1983; 0239 GMT; Coalinga, California; OFR 85-250 Records (13): Anticline Ridge: pad site Burnett Construction

Oil City

Oil Fields Fire Station: freefield and pad Palmer Avenue

Pleasant Valley Pump Plant: 1st floor, basement, roof, switchyard, freefield Skunk Hollow Transmitter Hill
TABLE A2. Processed records in each report. (continued)

April 24, 1984; Morgan Hill, California; OFR 84-498B, Vol. II. Records (11): Anderson Dam: downstream, crest Hollister City Hall Annex

Hollister Differential Array

San Justo Damsite: right abutment, left abutment

San Jose 101/280/680 bridge Hollister Differential Array No. 1, 3, 4, 5 December 23, 1985; with foreshock and aftershocks; Northwest Territories,

Canada; OFR 86- . Records (6): Nov. 9, 1985; 0446 GMT: Nahanni Site 2 Dec. 23,, 1985;; 0516 GMT: Nahanni Sites 1 Dec. 23,, 1985;; 0548 GMT: Nahanni Site 1 Dec. 25,, 1985;; 1543 GMT: Nahanni Site 3 January 26, 1986; Hollister, California; OFR 86-156 Records (5): Hollister Digital Differential Array, Stations 1, 3, 4, 5, 6

References

(1) Brady, A. G. , Etheredge, E. C., Maley, R. P., Mork, P. N., Silverstein, B. L., Johnson,

D. A., Acosta, A. V., Forshee, R. D., and Salsman, M. J. , 1986, Preliminary report on records from the USGS-maintained strong-motion

network in the Hollister area, January 26, 1986: USGS Open-File Report 86-156, 43 p. (2) Brady, A. G. , Mork, P. N., and Fletcher,

J. P., 1981, Processed accelerograms from Monticello Dam, South Carolina, 27 August

1978, and two later shocks: USGS Open-File

Report 81-0448, 35 p. (3) Brady, A. G., Mork, P. N., and Silverstein, B. L. , 1986, Processed strong-motion

records from the 2317:41, 2318:20 and 2318:40 aftershocks of the October 15, 1979, 2313:53

GMT earthquake; Imperial Valley. USGS Open

File-Report 86-441.

(4) Brady, A. G. , Mork, P. N., Perez, Virgilio, and Porter, L. D. , 1980, Processed

data from the Gilroy Array and Coyote Creek

records. Coyote Lake, California earthquake, 6 August 1979. USGS Open-File Report 81-42, 171

P-(5) Brady, A. G. , Perez, Virgilio, and Mork,

P. N., 1980. The Imperial Valley earthquake, October 15, 1979. Digitization and processing

of accelerograph records: USGS Open-File

Report 80-703, 309 p. (6) Brady, A. G. , and Perez, Virgilio, 1983, Processed accelerograms from Coyote Dam, California, March 25, 1978: USGS Open-File

Report 83-166, 82 p. (7) Brady, A. G., Porcella, R. L. , Bycroft, G. N., Etheredge, E. C. , Mork, P. N., Silverstein, B. , and Shakal, A. F. , 1984, Strong-motion results from the main shock of April 24, 1984, S. Hoose, compiler, in The

Morgan Hill, California earthquake of April

24, 1984, USGS Open-File Report 84-498A, v. I, p. 18-26, and v. II, 103 p. (8) , 1984. Strong-motion results from

the main shock of April 24, 1984, The Morgan Hill, California Earthquake of April

24, 1984, (S. Hoose, compiler) USGS Open File Report 84-498B, V. II, 103 p. (9) Maley, R. P., Etheredge, E. C., Johnson,

D. A., Switzer, J. C., Mork, P. N. , and Brady,

A. G., 1984, Strong-motion data recorded near Coalinga, California (May 2, 1983) and processed data from May 2 and May 9, 1983: USGS Open-File Report 84-626, 258 p. (10) Mork, P. N., and Brady, A. G., 1981, Processed accelerogram from Monticello Dam, Jenkinsville , South Carolina, 16 October 1979, 0706 Gmt: USGS Open-File Report 81-1214, 20 p. (11) Perez, Virgilio, 1982, The Imperial

Valley, California earthquake, October 15, 1979; time dependent response spectrum plots: USGS Open-File Report 82-183, 96 p. (12) Silverstein, Barry, 1985a, Processed strong-motion records from Monasavu Dam, Fiji; Earthquakes of February 13, 14, and 23, 1983: USGS Open-File Report 85-375, 58 p. (13) , 1985b, Processed strong-motion

records from the Coalinga, California

aftershock of July 9, 1983; 0740 GMT: USGS Open-File Report 85-584, 169 p. (14) Silverstein, B. , and Brady, A. G. , 1985, Processed strong-motion records from the Coalinga, California, aftershock of July 22, 1983, 0239 Gmt: USGS Open-File Report 85-250,

229 p. (15) Silverstein, B. L. , Brady, A. G., and 89
Mork, P, N. , 1986a, Processed strong-motion

records from Bougainville Island, Papua New

Guinea; earthquakes of December 13, 1981 and March 18, 1983: USGS Open-File Report 86-264,

148 p„ (16) 1986b, Processed strong-motion

records from the southern Alaska earthquake of

January 1, 1975; 0355 Gmt : USGS Onen-File

Report 86-191, 99 p. (17) Weichert, D. H., P. N, Mork, and others,

1986. The November and December, 1985, earthquakes in the Nahanni region of the Northwest Territories, Canada. USGS Onen-File

Report, in press.

90
DENSE INSTRUMENT ARRAY SYSTEM BY PWRI FOR

OBSERVING STRONG EARTHQUAKE MOTION

by

Toshlo Iwasakl, Yasushi Sasaki, Kellchl Tamura,

Koh Alzawa and Kazuyukl Takahashl

ABSTRACT

This paper presents the Dense Instrument

Array Program for observing strong

earthquake motions by the public Works Research Institute (PWRI), Ministry of Construction.

PWRI has Installed dense array

acce le rographs at Its Tsukuba campus In 1979. PWRI created a program to observe

strong motions at four sites; Sagara,

Yalzu, Numazu and Matsuzakl around

Suruga Bay, Slzuoka Prefecture. This research was completed by March, 1985.

At each site, 14-18 3-component

acce le r og raphs were deployed at 10-14 points at ground surface and In down-holes. By July 1985, records of 56 earthquakes were obtained at Tsukuba

campus; 11 earthquakes were observed at sites around Suruga Bay. This report Introduces the summary of

the PWRI's Dense Instrument Array system

with the preliminary analysis on amplification of ground motion and the velocity of horizontal propagation of

the ground motion by using the data

obtained at the Tsukuba campus.

1) Dr. Eng., Director, Earthquake

Disaster Prevention Department,

PWRI, MOC

2) Dr. Eng., Head, Ground Vibration

Division, Earthquake Disaster

Prevention Dept., PWRI, MOC

3) Research Engineer, Ground Vibration

Division, Earthquake Disaster Prevention Dept., PWRI, MOC

4) Ground Vibration Division,

Earthquake Disaster Prevention

Dept., PWRI, MOC

5) Kanto Regional Construction Bureau,

MOC

1 . INTRODUCTION

Based on knowledge accumulated through

past strong-motion observations. It Is widely recognized that characteristics

of ground motion differ, site by site,

and that characteristics of short period

ground motions are strongly dependent on seismic source mechanism, seismic wave path conditions from source area to observation station, and local geological and topographical conditions

near the station. To precisely

Investigate these effects, observation

through dense Instrumentation array, deployed In concentration and systematically In specific zones, is Indispensable, to currently adopted

conventional observation of strong motion . Considering this situation PWRI formulated a dense instrument array program to deploy a number of acce 1 e rog r a ph s at Its own campus In

91
Tsukuba Science City and at four local observatory sites In the Suruga Bay - Izu region. The latter was designated

as one of six high priority sites at the International Workshop on S t r ong- Mo t 1 o

Earthquake Instrument Arrays (Ref. 1). Fig. I shows the locations of the five sites, 1 .e . , Tsukuba, Sagara, Yalzu, Numazu and Matsuzakl. The observation

at Tsukuba campus was Initiated In 1979,

and deployment of arrays at Sagara,

Yalzu, Numazu and Matsuzakl was completed in March 1985.

The International Workshop on Strong-Motion Earthquake Instrument

Arrays classifies local effects arrays

into two categories, ground motion

arrays and special arrays. Ground motion arrays consisted of three subcategories; local laboratory

arrays, simple extended arrays, and elementary arrays and Special arrays consist of two categories;

so 1

1 - s t r uc t u re Interaction arrays and liquefaction arrays.

PWRI's array at Its Tsukuba campus Is a local laboratory array; Its arrays In the Suruga Bay - Izu region are simple extended arrays. The purpose of simple extended arrays Is to study the Influence of local geology and topography on the earthquake motions at different locations. To establish the criteria on response spectra for practical design applications It Is recommended to use three component

Instruments, measuring two horizontal

and one vertical components of acceleration .

2 . SIMPLE EXTENDED ARRAYS AROUND SURUGA

BAY - IZU REGION [Ref. 2-4]

In Fig. 1, the location of four local observatory sites around Suruga Bay - Izu region are shown. At these four sites, the digital type records of every

acce le rograph with absolute time are Independently acquired In each cartridge

recorder which has a pre-event memory of

5 seconds. Main specifications of the acce le rographs are listed In Table 1. Maximum range of acceleration Is 1.0 g and signals are digitized with time Interval of 1/200 second and are stored

on digital cartridge tapes. Time code generators having an automatic time corrector by NHK time signal are equipped for each recording system so that exact triggering time can be recorded at each station.

The deployments of array with subsoil conditions for each of the four sites

are described below.

2 . 1 Array at Sagara Site Sagara site Is located near Omaezakl in Shizuoka Prefecture facing Suruga Bay, approximately 170 km south-west of Tokyo

as shown In Fig. 1. Fig. 4 shows the array. Most of the surface of Sagara

site Is covered by soft alluvial clay

deposits. Baserock Is of the Sagara

Group, tertiary layers consists of alternating sandy and clayey rocks which

are common at the site; It has shear

wave velocity of approximately 70 m/s.

The thickness of the Sagara Group Is approximately 1000 meters.

Ten 3-component strong-motion

acce 1 e rog r aphs are Installed on the surface along an approximate L-shaped

configuration. Four 3-component

down-hole accelerometers are Installed

at four points as deep as 36-meters to obtain baserock motions. The signals of each down-hole accelerometer (same

specifications as those installed on the surface) are transmitted via cable to the recorders. Fourteen acce le rographs

we re Installed at 10 points at the Sagara site. The observation operation

started In February, 1982.

2 . 2 Array at Yalzu Site

The Yalzu site Is located near Yalzu

Port approximately 20 kilometers north

of the Sagara site. The array consists

of 12 ground-surface acce le rographs and

3 down-hole accelerometers as shown in Fig. 3. The left bank of the Seto river

Is covered by soft silty materials as deep as 110 meters. This bank has shear

wave velocity of about 100 m/ s . The right bank of the Seto river Is covered

by stiff gravel formations. It is expected that response characteristics

of subsurface ground between the banks

of the Seto river are very different.

The baserock at the site Is of the Takakusa Group; basalt In Neocene period, which exists at the ground

surface at observing point No. 1 and the depth gradually Increases In a direction

along observing point Nos. 4, 7, 8, 9,

10 and 12. The depth of baserock at the observing point No. 9 Is about 110 meters from the surface, and the deepest down-hole accelerometer is Installed at this level. The s t r o n g- mo t 1 o

acce 1 e rog raphs , the down-hole accelerometers and time code generators

used at this site are the same as those used at the Sagara site.

2.3 Array at Numazu Site

92
The Numazu site Is located In a western

part of Numazu city, Shizuoka

Prefecture. At the site, an array

deploys approximately north to south

from the foot of Ashltaka mountain to the coastal dune, crossing the low land

of Ukl shlmagahara . Grourrd surface at the district Is covered by organic soils

of 10 - 15 meters deep, underlayed by

gravel, sand and silt formation.

Baserock In this area is tuff breccia

and Inclines from north to south, shallow In north and deep in south. At the southern end of the array line, bedrock lies at the depth of 160 meters.

Shear wave velocity of surface soli ranges from 70 to 200 m/s and that of baserock Is approximately 650 m/ s

.

In 1980, PWRI initiated the deployment

of dense strong-motion observation at this site using the array of five

acce 1 e r og r a ph s [Ref. 5]. This time the site was called Ashltake site. Later,

the array at this site was reinforced

and the five acce le rographs were combined with new ones; the present

array was named Numazu site.

The array configuration at this site Is shown In Fig. 4. As shown In this figure, the array deploys 9 points along

the line from the foot of Ashltaka

mountain to coast dune and 3 points

along the line perpendicular to this deployment. Furthermore 2 more points

are settled at No. 11 and No. 12 points.

Total number of the points at this site

Is 14, and total number of

ac ce 1 e r og raph s Is 18 Including 5 down-hole accelerometers.

2 . 4 Array at Matsuzakl Site Matsuzakl site Is located at Matsuzakl

town on the south west coast of the Izu peninsula. The array deploys along the slope of Ushlbara mountain with stations

at the flat area, as shown In Fig. 5. The central part of Matsuzakl extends on

the alluvial fan made by the river Nakagawa and surrounded by mountains 200 meters high. The main line of stations

deploys from the low flat area of the alluvial deposit to the slope of the Ushlbara mountain In a north south direction. Five points are located on

the slope, an additional five points are located on the lowland along the main

line and one reference station Is deployed on lowland apart from the main

line. Fifteen ac ce 1 e r og r aph s are Installed at

11 points Including four down-hole accelerometers Installed at stations No.

5, 6, 8 and 9. The observation phase at this site started in March 1985.

3 . PRELIMINARY ANALYSIS PROGRAM FOR DENSE INSTRUMENT ARRAY RECORDS

Prior to analysis of records obtained at dense Instrument array sites, it Is necessary to check the characteristic

features of each data set which constitutes many components obtained by

14 - 18 accelerometers for one site during an earthquake event. To save

time for this preliminary analysis, PWRI developed a computer processing system

consisting of:

1) Graphic display of the eplcentral location with the location of four observation

sites.

2) Figure of the time history of recorded motion.

3) Figure of the calculated

response acceleration spectra

and Fourier spectra.

4) Display of the maximum

acceleration at each station.

5) Display of the duration time of observed ground motion at each station.

6) Display of the rate of maximum

acceleration at each station to one at an arbitrary station.

7) Display of the rate of response acceleration spectrum at each station to one at an arbitrary

station.

8) Display of the arriving

absolute time of ground motion

at each station . In Figs. 6-8, the examples of the output obtained by this system are shown

for the case of the records of the Nag an o ke n- Se 1 b u Earthquake of September

14, 1984, (M=6.8)

.

4 . LOCAL LABORATORY ARRAY AT PUBLIC

WORKS WORKS RESEARCH INSTITUTE CAMPUS

[Ref. 5] PWRI moved to Tsukuba Science City, Ibarakl Prefecture In March 1979, and Installed a local laboratory array for observing propagation of seismic ground

strain In vertical and horizontal directions.

93
PWRI Is located about 60 kilometers north-east of Tokyo and occupies 126 hectares (320 acres) of land. As shown

In Fig. 9, two field stations (A and B Fields) were selected for dense array observation.

Subsurface soil condition of the campus

Is comparatively uniform In the horizontal direction. The ground

consists of alternating clay, silt and sand with 100-meter depth; gravel and sand exist below the 100-meter depth.

Upper layers are alluvial deposits with thicknesses of 50 meters and lower

layers are diluvial. Shear wave velocities of the alluvial and diluvial

deposits are approximately 250 m/s and 400 m/s, respectively.

Fig. 10 schematically shows the deployment of array Instruments at Fields A and B. At Field A, thirteen 3-component acce le rographs are Installed

along a cross-shaped configuration with 100-meter limb length and with Instrumental arrangement consisting of three on the ground, five at 2-meter

depth, and five at 50-meter depth. At Field B, six 3-component ac ce 1 e r og r aph

are Installed along a L - Shape

configuration. The configuration has a 100-meter and a 50-meter leg;

arrangements are one at 2-meter depth,

four at 50-meter depth, and one at 96-meter depth. The direction of each

limb at Fields A and B Is north to south

or east to west, same as the direction

of each acce 1 e rog r aph component.

Characteristics of ac ce 1 e r og r a ph s are listed In Table 2. Signals from each

acce le rograph are transmitted by cables embedded In a duct to a central processing room.

The observation by this local laboratory

array was Initiated In July 1979. By July 1985, records were obtained during

56 earthquakes .

5 . ANALYSIS ON AMPLIFICATION OF MOTION

IN THE GROUND [Ref. 6]

5 . 1 Amplification of Maximum

Ac ce 1 e ration Amplification of maximum acceleration In the ground Is analyzed for the sets of records obtained at the 2 m and 50 m below ground surface (for example at points A2C0 and A46C0) at Field A of the Tsukuba campus . Records with relatively large

accelerations (larger than 10 gals

horizontally and 5 gals vertically)

obtained during 56 earthquakes with magnitudes of 4 or higher were analyzed.

The record sets analyzed are summarized

In Table 3. Fig. 11 shows an example of maximum

accelerations which were obtained at A2C0 and A46C0 during the event of EQ-28. The ratio of maximum

acceleration of the point at 50 m below

surface, to the maximum

acceleration at 2 m below the surface.

Is defined as follows.

n)

The value of for the case shown In Fig. 11 Is calculated as shown In Fig.

12. Average trend of ot for all analyzed

record Is shown by the mean value of

^ a and the standard deviation,

<7(2 as shown In Fig. 13. It Is known

from Fig. 13 that the value of a for

the records obtained at Tsukuba campus

Is 0.35-0.5 for horizontal component and

Is 0.4-0. 5 for vertical component.

Therefore, the amplification ratio In the ground of the maximum acceleration

Is 2-3 times for horizontal motion and

Is 2-2.5 times for vertical motion.

5 . 2 Amplification of Acceleration

Response Spectrum

An example of acceleration response spectrum Is shown In Fig. 14, which Is drawn for the records at A2C0 and A46C0 during the EQ-28 earthquake. The ratio

of the acceleration response spectrum

(5% damping to critical) of the record

at 50 m below the surface for each period, (T, 0.05) to the one

at 2 m below t ne ground surface for each

period, (T, 0.05) Is defined

as foil ows :

riT)

^QL-50m

( 0.05 ) ^QL-2m ( 0.05 ) ( 2 )

The value of y(T) for the case of Fig. 14 Is obtained as shown In Fig. 15. The mean value of y(T), ~((T) and the standard deviation Oy(T) for points A2C0 and A46C0 are shown In Fig. 16. Data from Fig. 16 suggests that;

94
1) The mean value of acceleration

response spectrum varies with

the period, but generally,

( T) In region of period less

than 1 sec. Is equal to 0.3-0.

6

(Amplification; 1.5-3) for horizontal components and 0.4-0. 7 (Amplification;

1.3-2. 5) for vertical components. This means that

the amplification In the horizontal direction Is slightly larger than that In the vertical direction In accordance with the amplification tendency of maximum acceleration. For the region with period more than

1 second, the ratio scatters

fairly wide; the mean value

In this region seems

to be bigger than that In the region with periods less than

1 second. This Implies that

the amplification for the component In the region with periods longer than 1 second

Is comparatively small.

2) The ratio of the acceleration

spectrum has the minimum value

of about 0.7 second In period.

Amplification Is maximum at this period. This Is In accordance with the estimated

predominant period of the ground equal to 0.79 second. Calculated from the shear wave velocities and the thickness of

the 1 aye r s .

6 . ANALYSIS ON PHASE VELOCITY OF GROUND

MOTION [Ref. 7) 6.1 Comparison between the Epicentral

Direction and the Predominant Direction

of Ground Motion propagation.

If It Is assumed that the ground motion

In horizontal plane propagates In a predominant direction, the cross correlation between the two components

of a recorded ground motion In this predominant direction and In the direction perpendicular to this direction Is thought to be minimum.

Prior to the calculation of the cross correlation, the two components of ground motion along two direction of an arbitrary direction and Its perpendicular direction are calculated

from the recorded data measured along north-south direction and east-west

direction. Assuming the east-west

component of the recorded ground motion

is expressed by and the north-south component Is expressed by

a/£), as shown In Fig. 17, the components of the ground motion along

the X axis and Y axis are rotated by

9 in anticlockwise, and dyKd.t) respectively, are expressed

as follows:

axis- £ ) = a,( t )cos d + fly ( £ ) sin ^ (3) ay(8, £ ) = -ttx(£)sin^ +Oy(£)cos^ (4)

The cross correlation coefficient of

a ( 5 • £ ) and ayid. t) , Rid)

Is obtained by the following equation:

ra^id. £) ayid. t)dt JC ax id. t)dt where, T denotes the duration time of the recorded data. As equations (3) and

(4) apparently leads tha t ^ t/=~a^ fdj t) and ayfd+TrjtJ=-ayidjtJ , then the relation ^(e-f-Tr)=R{e J is reduced. Fig.

18 shows an example of the results of R(©) reduced from the equation (5) by

varying every 10° from 0° to 180°, using

the recorded ground motions observed at the points of 2 m below the ground

surface (A2C0, A2N2, A2S2, A2E2, A2W2)

and at the points of 50 m below the ground surface (A46C0, A47N2, A45S2,

A47E2, A46W2) during the earthquakes of EQ-28 and EQ-34. In Fig. 18, epicentral

direction angle, dsp » which Is measured anticlockwise from the east direction. Is also Illustrated.

If the principal motion propagates In Its predominant direction, the maximum

value of the horizontal acceleration of the reduced component In this direction

Is thought to be smaller than that In the direction perpendicular to the propagating direction. Then, among the directions In which the cross correlation coefficient becomes zero, such direction In that the maximum acceleration Is smaller than the maximum

acceleration in the other direction, Is taken as the predominant direction of the ground motion propagation.

The predominant propagation directions

for each observing point and the epicentral directions are shown In Fig.

9. Although there are some earthquakes

where the predominant direction does not coincide with the epicentral direction.

In cases of three earthquakes, EQ-21,

EQ-34 and EQ-45, both directions

coincide well at points 2 m below ground surface. For earthquakes EQ-21, EQ-22,

95
[

EQ-34 and EQ-39, one sees that both directions coincide well at points 50 m below the ground surface.

6 . 2 Calculation of the Phase Velocity

of Ground Motion

Based on the results described above,

the horizontal phase velocity of the ground motion can be calculated for the five earthquakes, which the predominant

propagation directions coincide well with the epicentral directions. The

five earthquakes are EQ-21, EQ-22 (only

for the data obtained at points 50 m below the ground surface), EQ-34, EQ-39

(only for data obtained at points 50 m below the ground surface) and EQ-45

(only for data obtained at points 2 m below the ground surface). The horizontal phase velocities of the ground motion In epicentral direction

and perpendicular to the epicentral

direction between two points on a horizontal plane are respectively

calculated by dividing the horizontal

distance of these points by the phase

lags which make the cross correlation

coefficients of the ground motions at two points maximum In each direction.

The phase velocity In a vertical direction of the ground motion calculated from the phase differences of

the records between two points coincides

fairly well with the shear wave velocity.

As for the records obtained at two points 1 and j in Fig. 20, by denoting

the components at points 1 and j along X axis Which Is rotated by

(epicentral direction components) by

a{,x(Oand Oj.x(0 respectively, the cross correlation coefficient of these values, ^tj(r), Is expressed by the following equation.

( ' ) = 1% (6) J!, ‘i.xO>dt I

x),jU)di

where, means the phase lag. Let the phase lag at which becomes maximum value, R be denoted by “maz’and the projected distance between

two points to the epicentral direction

be expressed by i (refer to Fig. 20), then the phase velocity, V, Is expressed as follows.

The expression for the phase velocity of the component along Y-axis (component

perpendicular to the epicentral

direction) can be reduced similarly.

Table 5 summarizes the calculated

results by these procedures taking four sets of data at (A2N2, A2S2),(A2E2,

A2W2), (A47N2, A45S2) and (A47E), A46W2)

as the combination of points (1, j). In this table, the negative values mean

that the propagating direction is opposite to the apparent direction.

Although the shear wave velocity around

these points Is 250 m/sec for the layer

from ground surface to the depth of 50 m, and 400 m/sec for deeper layer, the phase velocity estimated by equation (7)

Is extremely large. Also there are many

cases where the propagating directions

are opposite to the apparent ones.

Next, the phase velocity of the treated records obtained by filtering the original records by band pass filter In frequency domain Is calculated for each

period by the same procedure mentioned

above . Fig. 21 shows an example of the results

of this calculation for the case of EQ-34 . It Is reduced from the analysis

for the phase velocity of various period

as follows:

1)

The apparent velocity IN horizontal direction of the ground motion In any period Is generally larger than the shear wave velocity.

2) There are no apparent relations between the period and the phase velocity as far as the range up to 5 second of period

which Is examined by this analysis. However, it seems

that the phase velocity

generally becomes smaller In longer periods.

3) It Is difficult to recognize

the general relation among the magnitudes of calculated phase

velocities from records obtained at the points 2 m below the ground surface and

the points 50 m below the ground surface . It is concluded from the above results,

that the apparent horizontal velocity of ground motion is extremely large for the data analyzed here because the ground motion propagates vertically around the ground surface.

[

[

[

[

[

[

c

[

c

i:[

[

[

[

[

L“[

96 C
7 . SUMMARY AND CONCLUSIONS

1) A dense Instrument array system was constructed and maintained by PWRI for observing strong ground motion during

earthquakes

.

2) The amplification factor of the motion In the ground of 50 m thickness

analyzed from the data observed at PWRI campus was 2-3 for horizontal motion and was 2-2.5 for vertical motion.

3) The phase velocity In a horizontal

plane calculated from the data observed

at PWRI campus was extremely large

compared with the shear wave velocity.

The reason Is considered that the ground

motion propagates vertically during the observed earthquakes.

8 . ACKNOWLEDGMENTS

The authors are grateful to the Ministry

of Construction, the Chubu Regional

Construction Bureau (Planning

Department, Shizuoka Highway

Construction Office, Numazu Construction

Office), Shizuoka Prefectural

Government, and the related municipal

Offices, for their cooperation for the installation of dense Instrument array

systems along the Suruga-Bay Area.

9 . REFERENCES

1) International Association for Earthquake Engineering: Strong-Motion

Earthquake Instrument Arrays, Proc. of

the International Workshop on Strong-Motion Earthquake Instrument

Arrays, Edited by Iwan, W. D., Honolulu,

Hawaii, U.S.A., May, 1978

2) Tadayoshl Okubo : On the Dense Array

Observation of Strong Earthquake Motion

(In Japanese), Civil Engineering

Journal, Vol. 22, No. 4, April, 1980

3) Toshio Iwasakl, Yasushl Sasaki, Kellchl Tamura, Koh Alzawa: Dense Instrument Array Program by Public Works Research Institute for Observing Strong

Ea r t hquake- Mo 1 1 on , ASCA Seminar,

Tsukuba, Japan, February, 1985

4) Yasushl Sasaki, Kellchl Tamura, Koh Alzawa, Kazuyukl Takahashl: Dense Instrument Array Observation by Public

Works Research Institute (in Japanese),

Civil Engineering Journal Vol. 28, No.

1 , January 1986

5) Tadayoshl Okubo, Toshio Iwasakl, Kazuhlko Kawashima: Dense Instrument

Array Program of the Public Works

Research Institute and Preliminary

Analysis of Some Records, U.S. -Japan

Panel on Wind and Seismic Effects, U.J.N.R., 13th Joint Meeting, May 1981

6) Yasushl Sasaki, Kellchl Tamura, Koh Alzawa, Kazuyukl Takahashl: Analysis on Strong Ground Motions during Earthquakes

observed by Dense Array at P.W.R.I. -- Study on Characteristic of amplification

and Spatial Variation of Ground Motion

-- (In Japanese), Technical Memorandum

of Public Works Research Institute, No. 2270, December, 1985

7) Yasushl Sasaki, Kellchl Tamura, Koh Alzawa, Kazuyukl Takahashl: Analysis on Strong Ground Motions during Earthquakes

observed by Dense Array of P.W.R.I. -- Study on Apparent Velocity of Wave Propagation -- (In Japanese), Technical Memorandum of Public Works Research

Institute, No. 2271, December 1985

97
Table 1 Main Specifications of Accelerograph used for Simple Extended Arrays around Suruga Bay

Acceieromecer

Type Triaxial, Velocity Feed-Back

with Calibration Coil Full Scale Range 1 i IG Natural Frequency 5 Hz Damping Ratio >30

Trigger System

Sensitive Direction I Vertical Acceleration Set Point 1 0.005G (Changeable) Operation Cycle

[ Self-actuating for duration of earthquake, with automatic reset

i 5 seconds after averaged accelI eracion in horizontal direction

1 becomes less chan 5 gals

Recording System

Type j Digital Number of Truck 1 ^ Magnetic Tape

1 Digital Cartridge; 60 inch/sec. Number of Bits 1 16 bits

I Dynamic Range 1 r 92 dB Sampling Race

1 200 Samples/sec/ channel

Max. Recording Time

I Approximately 40 minutes pre-event Memory 5 seconds for each channel Time Code Generator

Accuracy of Crystal i 10-'^ Auto Adjustment of Time of Crystal

Each 1 hour by NHK Time

Signal

Power Requirement

Voltage 12 VDC

Battery Charger Float Charger Supplied; Approved

for less chan 25 hours after suspencion of commercial electricity

Case Water tight case for proof of twice of atmospheric pressure

Table 2 Main Specifications of Accelerometer used

by Local Laboratory Array at PWRI

Number of Component | 3 Type Velocity Feed Back Natural Frequency j 5 Hz Frequency Range j 0.1 - 50 Hz ; within 1 dB Maximum Acceleration

j + 650 gals Diameter, Length

| i 13 cm. 125 cm Weight 50 kg
Table 3 Number oi' Records used for Analysis on Amplification

of the Motion in the Ground

CoBponent N-S E -W U-D

Co«btn»tlon'\. A4 6C0

3 0 A2C0 2 5 2 9 A4 7N2 A2N2 3 0 2 5 3 0 A45S2

3 1 A2S2 3 0 3 0 A47E2

3 0 2 7 2 9 A2E2 A46W2

2 3 A2W2 1 9 2 5

Table 4 Earthquake Records used for Analysis on Wave Propagation

Eftrtbquaic* Data Hypoc.nt.r Epic antral Distance

Ckm]

JMA

Ho. Rsglon i Longitude Utltud. Depth 0dn3 Magnitude EQ - 13 1 9 80. 9. 25 Central' CMba Pref. 14 0°1 3' 35°3 1' 80 69 6.1 EQ - 16 19 8 1. 9. 2 Off Ibaraicl Pref, 1 4

1 °08' 35°48' 4 0 102 5.8 EQ - 21 1 98 2. 3. 7 Kaahiiia«nada 1 4 0“39' 36°28' 60 64 5.5 EQ - 22 1 98 2. 7. 23 Off Ibwdtl Fraf. 1 4

1 “5 7' 36°1 1

' 30 1 6 9 7 0 EQ - 28 1 98 3. 2. 27 S.TbaraJcl Pref. 140°09' SS'^SS ' 72 22 5.0 EQ - 31 1 98 3. 7. 2 Off FuJcuahlua Pref. 1 4

1 ®07' 36 "Sd

50 127 5.8 EQ - 34 1 9 83. 1 0. 28 SE tbarakl Pref. 139°59' 36°1 3' 60 1 3 5.2 EQ - 39 1 9 84. 1. 1 3 off ginitl

1 36°59' 33°1 6'

4 00 4 25 7.4 EQ - 44 1 9 84. 2. 21 SI Ibmkl Pr.f 140“06' 36°06' 70 3 5.2 EQ - 45 1 9 84. 3. S Hmt TorlsMM 1 39“’08' 29*28

460 7 4 5 7.9

99
Table 5 Apparent Velocity of Wave Propagation

Earthquake

Mo.

Epicentral

Direction

d EP

Predooinant

Direction

P Observing

Points

1*opagatlon Distance ^Cm]

Coopo Hax. Cross Correlation

Phase Lag r.ax Csec]

Propagation^ ^ Velocity

K Cm/sec]

EQ - 21 4 2,6 40

A2N2 — A2S2 64.3 LG 0.5 7 7 0.0 1 1 6 4 3 0 TR 0.6 1 3 0.0 0 OO A47N2— A45S2 64.3 LG 0.3 7 4 - 0.0 5 [ - 1 286

T R 0.4 9 7 - 0.0 5 - 1 2 86 A47E2— A46W2 7 6.6 LG 0.5 3 6 0.0 1 7 6 6 0 TR 0.7 5 3 0.0 0 OO E Q - 22 2.8 0 A47E2 — A46W2 1 0 0.0 LG 0.7 9 6 0.0 0 GO T R 0.8 3 4 0.0 0 OO

EQ - 34 12 2.3 1 2 0

A2N2 — A232 866 LG 0.3 9 8 0.0 0 a= TR 0.6 5 7 -0.0 2 - 4 3 3 0 A2W2 ^A2E2 5 0.0 LG 0.2 0 9 0.0 5 1000 TR 0.5 2 8 0.0 3 5 , 14 2 9 A47N2— A45S2 8 6.6 LG 0.2 6 1 0.5 2 ! 167

T R 1 0.5 0 6 0.0 1 3 6 6 0 A46W2 — A47E2 5 0.0 LG 0.4 9 7 -0.0 2 - 2 5 0 0 TR 0.6 2 3 -0.0 2 - 2 5 0 0

E Q - 39 2283 2 30 A45S2-.A47N2 7 6.6 LG 0.4 7 7 -0.0 1 - 7 6 6 0 TR 0.6 0 9 -0.0 2 - 3 8 3 0 A46W2 — A47E2 6 4.3 LG 0.6 7 0 -0.0 1 - 6 4 28 TR 0.7 2 4 -0.0 1 - 6 4 2 3 EQ - 45 2 6 4.3 270 A2S2 -A2N2 10 0.0 LG 0.8 5 0 0.0 0

1 T R 0.8 7 7 0.0 0 I

1) LG: Epicentral direction TR; Perpendicular to LG

2) Negative velocity means that the propagating direction is opposite to the apparent direction.

SIMPLE EXTENDED ARRAYS AROUND SURUGA BAY-IZU REGION

No LOCATION

YEAR OF COMPLETITION © SAGARA 1982

YAIZU 1983 © NUMAZU 1984

MAT5UZAKI • 1985

TSUCHIURA CITY ^KASUMIGA-URA X7~

Figure 1 Location of Dense Instrument Arrays by PWRI

100

FiCTj,re L Array ar Numazu Site

Figure 5 Array at Matsuzaki Site

102
1 1 ^ ! TP .4

{ r4 4

^4v^'

S'

r'

) J

4 +

'i, Aj-A<> C4

ir'~

7t.

j

!> 0

0

i3T.« Figure 6 Location of Epicenter (Nagano-Ken-Seibu

Earthquake , 1 984

)

Figure 7 Maximum Acceleration of Yaizu Site

103
Figijre 8 Maximum Acceleration of 4. Sites around Suruga Bay

Figure 9 Local Laboratory Array at PVRI

104
iLoanoN (N.s.E.wi CEPTH FROM SURBCg [ml IFIELD-A

Note AOE1 : Observation has been stopped since September

16, 1982.

Field B

Figure 10 Array Configuration at P/VRI Campus

105
MAX. ACCELERATION CgoJ

Figure 11 Example of Maximum Acceleration (EQ-28)

MAX ACCELERATION RATIO

Figure 12 Example of Maximum Acceleration Ratio (EQ-28)

(a) N-S Component

(b) E-W Component

(c) U-D Component

Figure 1 3 Average and Standard Deviation of

I-laximum Acceleration Ratio

106
100

Figure 14 Example of Acceleration

Response Spectrum (EQ-28

)

1 ill! !

‘

i M !l TT^

EQ-28 A46C0/A2C0

• N-S

. E-W A- 0.05 U-D

/. / -IK,

j

1

1

1 .

i a.J 0.3 O.S 0.7 I 3 3 5 NflTunPt. PERIOD TI5ECI

.Ji46C0 A3C0 .S-3

7 /i=0.05

•|n!|

i

i iii

il

'i

i

,Si ,

i 1 J.l!|j]j^

rlJj

!

!

os o.l 0.2 0.3 In

0.3 0.

l!

1 1

1

1 . Lll 2 3

(a) N-S Component

(b) E-W Component

(c) U-D Component

Figure 15 Example of Acceleration

Response Spectrum Ratio (EQ-28) Figure 16 Average and Standard

Deviation of Acceleration

Response Spectrum Ratio

(A46C0/A2C0)

107
CROSS

CORRELATION

COEFFICIENT

R

^^O^S

CORRELATION

COEFFICIENT

R

Figure 17 Rotation of Coordinate

20 40 60 80 100 120 140 160 180 ANGLE OF ROTATION 9 (DEGREE)

20 40 60 80 100 120 140 160 180 ANGLE OF ROTATION 9 (DEGREE)

(a) EQ-28, GL-2m (b) EQ-28, GL-50m

ANGLE OF ROTATION 9 (DEGREE) ANGLE OF ROTATION 9 (DEGREE)

(c) EQ-3^, GL-2in (d) EQ-S**, GL-50m

Figure l8 Relation between Rotation Angle and Cross Correlation Coefficient

108
v /55° v/s6 ^'sA46co

/ A . . /so- 9ep= 284.5 6L-50m

A45S^

(b) EQ-28, GL-50m

(d) EQ-3^, GL- 5 O1H Figure 19 Predominant Propagation Direction

Figure 20 Definition of Propagation Distance

109
VELOCITY

OF

WAVE

PROPAGATION

v

tm/sec)

VELOCITY

OF

WAVE

PROPAGATION

v

tm/sjc)

2 3 4 PERIOD TCsec] (c) EQ-34, Al+TH2-^Al+5S2

1 2 3 4 5 PERIOD T[sec) (b) EQ-34, A2W2^A2E2

>. -4000 o

LG

TR 3 -6000 >

LGTR 12 3 4 5 PERIOD T[sec)

(d) EQ-3^, A46W2-A47E2

Note LG: Epicentral direction TR: Perpendicular to LG

Figure 21 Propagation Velocity of Various Period

no
VIBRATING TESTS OF AN ASPHALTIC FACING COFFERDAM BY BLASTINGS

BY

N. Matsumoto^

N. Yasuda

M. Shiga

ABSTRACT

Rockfill dams with upstream impervious facings

have advantages in rapid construction and economical construction cost. However, impervious

facings are vulnerable to damage caused by the settlement of the dam and earthquake motion.

Therefore, dams of this type have pervious rock

zones capable of draining leakage, if any, and adequate means of lowering water level for repairing the impervious facing. On the other hand, compaction of the rockfill by vibratory roller and use of well graded materials enabled minimization of the post construction settlement. Vulnerability of the facings

of the rolled fill dam decreased compared to the dumped fill. Dams with facing often have been

designed in areas where seismic activity is low. The seismic resistance design of this type of the dam is important to Japan with its frequency

of strong earthquakes. So the vibration test

test of a 16.5-m-high asphaltic facing dam by using blastings at the quarry site was performed. Seismographs, accelerations and dynamic

strain gauges were placed in the asphaltic

facing dam and its dynamic behavior was investigated. The quarry site was exploded when the reservoir level of the dam was at its lowest, intermediate, and the highest. Accelerations

recorded at the dam foundation ranged from 4 gals to 30 gals.

In this paper, the effects of the reservoir

level to natural frequency and response amplitude at the dam crest are reported. Also, the behavior of an asphaltic facing during vibrating

and the comparison with calculated results by a computer program are discussed.

1. GENERAL DESCRIPTION OF A COFFER DAM AND BLASTINGS

The 110-m high rockfill dam (central core zone

type) is now under construction by Hokuriku

Regional Construction Bureau of the Ministry of Construction. A 16.5-m-high asphaltic facing

dam was constructed as the upstream No. 1 cofferdam, the downstream slope was covered with

concrete

.

In this cofferdam seismographs, accelerometes

,

and strain gauges were installed to monitor

dynamic behaviors during explosions. Seven seismographs were located in the dam. Four were

inside the dambody and the others were at the

surface of the dam. Seismographs (made by Katsujima Co., Ltd.) were servo-type strong motion accelerometers which were composed of three components (up-down stream, dam axis, and vertical directions); 21 components of earth- quake motion were observed. Figure-1 shows the positions of the seismographs. Accelerometers (made by Kyowa Electric Co., Ltd) were strain gauge type. Each meter was composed

of up-down stream and vertical direction, and 12 components of earthquake motion were obtained. Figure-2 shows the locations of accelerations

and strain gauges. The vibrating tests by blastings were performed at the upstream No.-l cofferdam (Asphaltic facing dam) and the response to the shaking of a cofferdam was investigated.

The blastings were conducted three times corre- sponding to three different water levels, i.e.

at the water level before filling the reservoir,

at the top water level and at the intermediate water level (twice). Table-1 shows the list of blasting tests. The locations of explosions

were at the downstream side of a quarry site

(400 m in the horizontal distance) during the time of an intermediate water level and the downstream side of a quarry site (400 m in the horizontal distance) for the others. Powder quantity used were 1,004 kgf—1,715 kgf for each

case.

2. BEHAVIORS OF A COFFERDAM DURING BLASTINGS

Observed waves were digitized at every 0.002

seconds. The maximum accelerations, the maximum

displacements and power spectra were obtained

for each explosion wave. Each spectrum was smoothed by 0.4Hz of Parzen's spectral window. Figures 3 & 4 show time histories of accelerations and spectra at the crest and foundation.

The maximum accelerations, the maximum displacements and predominant frequencies observed by each seismographs are indicated in Tables 2 & 3. The maximum acceleration of the foundation in the up-down stream direction ranges from 4.7 gals to 20.1 gals; for the dam crest it ranges

from 16.4 gals to 67.7 gals. The response amplitude at crest to foundation is about 3.5. The dynamic behaviors of asphaltic facing are shown by Figure 5. Time histories are presented

of displacement gained by integration of accel- erations which were observed at V-2 (the surface

111
of an asphaltic facing) and V-7 (the inside of a facing)

.

The relative displacement between an asphaltic

facing and the dam body is very small, as are

the phase difference between them. The behaviors

of a facing appear to be almost the same with

that of the dam body. This is distinct from the power spectra of V-2 and V-7 both in the up-down

stream direction because their spectra have similar shapes when the reservoir is empty and

full. Table-4 shows the maximum accelerations

observed by accelerometers and the maximum

strains by strain gauges, both types are located

at the asphaltic facing. In the case the maximum acceleration at the foundation is more than

10 gals, the maximum in up-down stream direction

tend to be larger than those of vertical direction. The maximum strains range from

1 X 10~^ to 1 X 10“^ and likely to decrease in lower level.

The wave velocities at the facing and the dam

body were 122-167m/sec and 341-441 m/sec respectively during the empty reservoir, and 122-192 m/sec and 341-375 m/sec respectively

during the full reservoir. In disregard of the path of wave propagations in the dam body, the wave velocities were calculated from distances

between seismographs and start of waves. The propagation speed at an asphaltic facing was the half of the dam body. Both velocities decreased

after the reservoir was filled. The natural frequency of the dam body can be estimated from the power spectrum of acceleration observed at the dam crest, but the fre- quency characteristics of foundation has to be eliminated from this spectrum. The time history

of acceleration of the foundation and the dam

crest were assumed to be f(t) and g(t) respectively. Fourier spectra F(w) and G(w) are obtained as follows: F(w>- (1)

frequency response of the reservoir function

during filling. During filling, the first natural frequency becomes smaller and the response amplitude decreases from 16 to 10. The authors reported test results of a two-dimensional dam model by a shaking table during the sixteenth (1984) joint meeting of UJNR. This result coincided with the above-mentioned one

i,e., the response amplitude and the natural

_

frequency decreases as it is being filled (see Figure-8)

.

3. DYNAMIC ANALYSIS OF AN ASPHALTIC FACING DAM

Using the time histories of acceleration of the foundation during blastings, the observed

response accelerations of the dam body are com- pared with computed ones. The computer program

used was DINAS; modal damping method was used. The part of kinetic energy accumulated in each mode dissipated as damping energy. At first, kinetic energy of each element is calculated

from the mass matrix of each element for every

natural frequency of free vibration. Damping

ratio of each natural mode is given as follows:

hi- 2 Cjx'mjXj

j=i^ ^.nijXi

j=i > '

' X' MXi

mj : Mass matrix of each element j Cj : Damping ratio of element j M : Mass matrix of structure

xi : i - th mode shape

(4)

The Westergaard' s added mass concept is adopted

for the hydro-dynamic effect of the upstream

face. The fiildam body may be considered elastic when very small strain is expected to occur. The maximum strain observed at the dam

body was 1 x 10“^ during blastings, so two- dimensional elastic analysis was performed as in

the concrete dam. Figure-9 shows the finite element idealization. G(w)- (2, The frequency response function of the dam crest

to the foundation Zw is given by the equation: Z (w}= G (w) / F (w) Upstream-No. 1 is a three-dimensional structure,

the vertical motion of the foundation, for example, contributes to the vertical and to the horizontal motions of the dam. However, ignoring this coupling motion, the frequency

response functions of up-down stream component,

shown in Figure-7 , were obtained from equation

(3). From this response function, the first natural frequency of the dam is 8.1 Hz when the reservoir is empty. Figure-7 also shows the

It is known that shear modulus of rock materials

depends on the confining pressure. However, the modulus of the dam body was assumed to be inde- pendent of the pressure and 2,997 kgf/cm^ of modulus was adopted in this computation. Poisson's ratio and damping ratio is 0.36 and

10%, respectively. The input properties used

for computation were obtained after the equations of Sawada et al. The characteristics of

the asphaltic facing were determined from lab- oratory tests as shown in Table-6. The calculated and observed results are indicated in Table-7. Table-8 shows the maximum computed

strain of an asphaltic facing. Figures-10 to 12 show the time histories of acceleration and dis- placements and frequency response functions during the empty reservoir (case-1) and full (case-2). There were coincidences between the computed result and the observed data, especially

about the natural frequency and the accelerations

112
of the dam body.

4. CONCLUSION

The following are results from investigations of dynamic behaviors of the asphaltic facing dam.

1. The response amplitude of the dam crest to the foundation is about 3.5.

2. Relative displacements between the asphaltic facing and the dam body was very

small, and the phase difference was difficult to recognize; their dynamic behaviors

during blasting were near similar to each

other.

3. The natural frequencies and the response

amplitude of the dam body decreased as the reservoir was filled. The same results

were obtained from a two-dimensional dam

model using a shaking table.

4. The computed vertical accelerations of the dam crest were nearly equal to the observed

ones in every case. The response dis- placements and frequencies were the same

in vertical and horizontal directions. ACKNOWLEDGEMENTS

The authors appreciate the assistance from members of Sagurigawa Dam Construction Office of Hokuriku Regional Construction Bureau. REFERENCES

(1) N. Matsumoto, M. Toyoda, M. Shiga, "Dynamic

Tests of Rockfill Dam Models," 16th UJNR,

1982.

(2) "Dynamic Interaction Problem Analysis

System," C.R.C. Co., LTD.

(3) Y. Sawada, T. Takahashi, A. Sakurai and

H. Yajima, "Material Property Distribution

and Dynamic Characteristics of Rockfill

Dams—A Consideration Based on Elastic

Wave— ," Central Research Institute of Electric Power, Report No. 377008, 1977

( in Japanese)

.

(4) N. Matsumoto, N. Yasuda, M. Shiga, " Measured Behaviors of Asphaltic Concrete

Facing Dam," 1986 (unpublished).

113
Table 1 General Description of blastings

Water

level (m)

Temperature of Asphalt Location of Blastings (m)

Horizontal

Distance (m)

Surface 10 cm

below EMPTY empty 38.0 41.0 Upstream side of Quarry EL = 420

200 FULL EL =

368.22

36.5 33.0 Upstream side of Quarry EL = 400 ~ 420

200 INTERMEDIATE-1 EL = 364.01

34.0 32.0 Downstream side of Quarry EL = 550

400 INTERMEDIATE-2 EL = 362.33

36.0 38.0 Downstream side of Quarry EL = 550 400

I A-A Section

7.000

114
Table 2 The maximum acceleration and displacement in each explosion — Empty Full Intermediate1

Intermediate-2 Seismograph Direction

Acceleration

gal

Displace- ment cmAcceleration

gal

Displace- ment cm

Acceleration

gal

Displace- ment cm

Acceleration

gal

Displace- ment cm

V-1 Up-down stream 46.4 0.016 67.7 0.016 16.4 0.008 27.3 0.016 Dam axis 67.3 0.022 65.9 0.016 16.0 0.008 25.3 0.013

Vertical 48.6 0.009 68.5 0.015 8.0 0 002 14.2 0.003

V-2 Up-down stream 36.2 0.006 32.5 0.006 10.3 0.004 14.8 0.008 Dam axis 40.3 0.009 33.1 0.006 6.4 0.003 10.7 0.005

Vertical 55.0 0.006 48.4 0.001 8.9 0.001 20.6 0.003

V-3 Up-down stream 32.0 0.009 63.9 0.011 9.3 0.004 20.0 0.009 Dam axis 67.8 0.012 66.7 0.011 9.3 0.004 8.8 0.005

Vertical 54.3 0.008 106.2 0.009 8.1 0.001 20.3 0.003

V-4 Up-down stream 39.2 0.015 41.6 0.016 14.9 0.008 26.4 0.015 Dam axis 69.8 0.021 62.7 0.016 15.1 0.007 25.4 0.013

Vertical 39.1 0.010 56.8 0.015 6.4 0.002 14.1 0.003

V-5 Up-down stream 21.8 0.010 26.1 0.011 11.4 0.006 18.6 0.011 Dam axis 43.0 0.014 44.7 0.013 9.3 0.005 21.1 0.010

Vertical 35.0 0.009 58.9 0.013 5.9 0.002 8.8 0.003 V-6 Up-down stream 15.5 0.004 20.1 0.004 4.7 0.003 8.4 0.005 Dam axis 20.7 0.005 24.0 0.006 4.0 0.002 8.5 0.004

Vertical 24.8 0.005 50.4 0.009 4.3 0.001 9.7 0.003

. V-7 Up-down stream 32.0 0.007 32.9 0.008 9.6 0.004 15.1 0.008 Dam axis 35.3 0.009 31.6 0.006 6.6 0.003 12.0 0.005

Vertical 42.9 0.007 64.5 0.012 8.8 0.001 21.6 0.003

Table 3 The predominant frequencies of observed accelerations in each explosion

Predominant frequencies Hz

Seismosreph Direction Empty Full Intermediate1

lntermediate-2 X

6.3 18.0

8,0 19614.4 17.5 lO^ 12 4 14.4 6 6 11,1 6 3 V-l Y

6.9

14.8

8.0 11.1 12.5 6 9 8.1 10 8 14.7 6.6 82 9 2 6 5 8.0 z

9.1 18.7

12.8 15.9 17.3 9 1 12.9 15 3 8.8 11 9

5 5

18 5 8.921.1

14.0 IT.T X

6 3 19.3 9 3

20.9

14.0 24 0

15.9 6.3 8 4 9 3 14 0

6 0

10 9

6.6

13 4 8.517 7 9 5

! 13 2 17 5 v-2 Y

6 8 18 511.1 22,0

12.0 30 015.2 6.8 17 79.3 22.412.0 30.0

14.8 60

10 9

6,6 8.2 10 0 r 4 6.6 8 0 20 011 2 z

8.8 20 3

12.7 21,514.8 29 6 19 1 8 7 22 3 12.8 15.5 20.8

8.5 11,

4

18.2 ! 20 0

11,4 15 9 180 X

6.4 25 3 12.0

27.1

18.5 28.5 21 6 6 4 15 3 10.4

17.3 12.027.6

13.9 6 8

14 6

8 0

17 8

8 9

18 3 20,0

§-? 8,2 10 6 143

V-3 Y

6.5 17.5 10 4

18 8 12.8 16 3 6 5 8.8 12.3 17,6 5 8 7 5 9.5 14 2 5.9

' 13 1 6 6

14 9

9 2 z

8.9 26 8

15.1 19.6 21 3 8.7 25.3

14.0 30.516.9 42.0 20.0 8 6 22.611.7 17.5 21.8 ! 8 3

! 19 5

12 0 14 3 17 9 X

6.3 18,0 10.3 14.0 15.5

6.4 10.4 14.0 6.6 10,9 6.3 v-4 Y

6.8 10.8

6.8

14.7 8.0 10.8 12.5

6 5 8.0 9 2 . J 6.6 8 0 z

9.1 13.1 17.6 20.3 9.1 13.1 6 4 8.9 I 8 9 13 5 X

6.4

18 0

8 7 lO.O 12.5 6.4 10 0 17.5 6.6 1 6 0

V-5 Y

6.8 18,0

10.8 12.5 16,5 69

15.5 8.0 17.6

10.8 12.7 6 0 6J 8.2 92 6.5 8 0 z

9.1 13.1 18.0 20 0 9.1 13.3 8.8 8 8 13 5 X

6.1 20 8 10 1 14.0 17 3 6 3 16 8 8 3 19.3 10 9

33.1

13.7 35 1 6 0 8 3 10 8

1 6.0 V-6 Y

6.5 13 6 18.7 26 0 6 8 10 8 14 0 18 0

1 6 0 5.9 6 6 Z

8.8 18 0 19 3 8 8 8 8 20 0 X

6 1 10.0 14.0 6 3 9 9 14.0 19 1 8.613 2 6.6 17 8

8.3 9 5

6 0 17 5 V-7 Y

6 8 23 611.1 12.3 15 1 6 9 17 39 3 22 3 120 14.8 1^96.6 8 2 10 0 66 n I 15 2 Z

8 8 12.5 14 9 23 2 8 8 12.7 15.6

8.5

18 5 18 9

6.) ITS

8 6 1.3 16.0 18 2 Not*) Thick numbers mean Che majiimum power spectra

115
EL

--370

--365

--360

--355

Figure- 2 Positions of accelerometers and strain gauges

Table-4 The maximum acceleration (by accelerometers! and the maximum strain (by strain gauges)

in each explosion — Empty Full Intermediate1

Intermediate-2

Accelero- meter

Direction

Acceleration

gaJ

Strain xl0‘‘Acceleration

gal

Strain

X 10'*

Acceleration

gal

Strain

X 10‘‘

Acceleration

gal

Strain

X 10'* A-1 Up-down stream 39.3 1.448 56.5 1.272 16.6 0,680 25.4 0.712

(S-1) Vertical 47.7 2.4252) 68.3 1.817 11.6 0.668 15.4 0.775

A2Up-down stream 31.5 2.207 •> 34.4 2.094 12.9 0.655 20.8 1.048

(S-2) Vertical 56.7 1.228 2) 83.5 0.802 14.3 0.805 19.8 0.807

A-3 Up-down stream 37.0 2.708 ') 50.8 2.078 11.4 0.850 17.6 1.368

(S-3) Vertical 38.1 1.857 2) 54.3 1.110 8.4 1.011 19.1 0.883

A-4 Up-down stream 39.9 0.841 ') 36.0 0.863 8.1 0,760 17.1 0.732

(S-4) Vertical 28.7 1.068 2) 31.7 1.929 8.4 0.672 12.7 0.827

A-5 Up-down stream 30.6 0.963 ') 24.9 0.791 4.3 0,668 7.6 0.770

(S-5) Vertical 39.3 0.977 2) 27.1 0.831 4.3 0.699 10.4 0.745

A-6 Up-down stream 16.2 0.907 ') 24.9 0.848 4.0 0.707 6,7 0.757

(S-6) Vertical 25,5 0.870 2) 26.0 0.726 3.2 0,740 9.4 0.886

1) Strain on the surface 2) Strain of the inside

116
Vertical (V-6)

Vertical (V-1)

Figure-3(a) Time histories of accelerations at the foundation

(V-6) and dam crest (V-1) during the empty

reservoir - 8 -

117
Up-down stream (V-6)

I

Dam axis (V-6)

V-S-Z

Vertical (V-6)

Vertical (V-1)

Figure-3(b) Time histories of accelerations at the foundation

(V-6) and dam crest (V-1) during the full reservoir

118
L.'«hm Tj-t nju fuu

Up-down stream (V-6)

OkiM l-x

Vertical (V-6) Vertical (V-1)

Figure-4 Power spectra of accelerations at the foundation

(V-6) and dam crest (V-1) during the full reservoir

119
DISPLACEMENT

(cm)

DISPLACEMENT

(cm)

DISPLACEMENT

(cm)

DISPLACEMENT

(cm) 0.010 A /V ^

0.00 10.50 l.OO 1.50 2.00 2.50 3.00TIME

3.50 4.00 4 IN SECONDS Empty reservoir (V-2).AM/

0.00 10.50 1 . 00 1.50 2.00 2.50 3.00TIME3.50 4.00 4 IN SECONDS Empty reserovir (V-7)

Figure-5(a) Time histories of displacements at V-2 and V-7

during the empty reservoir

TIME IN SECONDS

Full reservoir (V-7)

Figure-6(b) Time histories of displacements at V-2 and V-7

during the full reservoir

,50

,50

50

50

120
li1I

121

Figure-6

Power-spectra

of

vertical

accelerations

at

V-2

and

V-7

during

the

empty

and

full

reservoir
FHKQUNKCY

HESFONSfe:

FUNCTION

LegendEmpty

Full Intermediate1

lntermediate-2 Note) 0.6 times of scale is during empty

A

j \ .y \ Ah ^ J

A \ / \

8.00 10.00 12.( FREQUENCY iCPS)

U.OO 16.00 18.00 20.(

Figure-7 Frequency response function of up-down stream

direction at the dam crest in each explosion

10

FREQUENCY (Hs) Figure-8(b) Comparison of the frequency response functions of two-dimensional dam model during the full cind empty reservoir

122
Table-5 Computed cases CASE Water Level Input Wave

Evaluation of Hydro-dynamic

Force

1 Empty Vertical and Horizontal

2 Full Vertical and Horizontal Westergaard 3 Intermediate1

Vertical and Horizontal Westergaard

4 Intermediate2

Vertical and Horizontal Westergaard

Table-6 Inputed properties of materials

Asphalt Rock

Unit Weight (t/m^) 2.36 2.1

Elastic Modulus (t/m^

)

5500 70730

Poisson’s Ratio 0.40 0.36 Damping Ratio (%) 10 10

NP.55

123
124

Figure-10

Computed

accelerations

of

the

dam

crest
125

Figure-11

Computed

displacements

of

the

dam

crest
Frequency response function at N.P.55 during empty (Case 1) r~ GRID 55 - X

1

1

1 1

1

i

1

l\

1 A /I 1

00 U=

r

s J

1 =J

f»€Out»C» <H.» Frequency response function at N.P.55 during full (Case 2)

Figure-12 Frequency response function of the dam crest

126
Table-7 Observed and computed results

Case

Natural Frequency (Hz) Maximum Acceleration (cm/s^) Maximum Displacement (cm)

1st 2nd 3rd Horizontal Vertical Horizontal Vertical

Empty 7.11 11.19 12.91 -26.6 45.8 -0.013 -0.009 (CASE-1) (8.0) (9.0) (13.0) (46.4) (48.6) (0.016) (0.009)

Full 6.94 10.84 12.81 -32.7 68.7 -0.015 -0.014 (CASE-2) (7.7) (9.1) (12.2) (67.7) (68.5) (0.016) (0.015)

Intermediate-1 7.08 11.10 12.89 18.5 10.1 0.009 -0.002 (CASE-3) (7.1) (9.0) (12.4) (16.4) (8.0) (0,008) (0.002)

Intermediate-2 7.10 11.16 12.90 32.5 -14.1 0.017 0.003 (CASE-4) (7.1) (9.0) (16.4) (27.3) (14.2) (0.016) (0.003)

Note) ( ) is the observed value

Table-8 The computed maximum strains in an asphaltic facing \^ase Empty Full Intermediate1

Intermediate2

Element'-''-^ (CASE-1) (CASE-2) (CASE-3) (CASE-4)

61 1.82^-6 4.12^® 0.394®^’® 1.15®-® 62 1.66 3.59 0.357 1.00 63 0.822 1.41 -0.312 -0.608 64 3.03 -6.70 -1.18 -1.96 65 -3.39 -13.3 -1.66 -3.11 66 -3.01 -26.8 -1.43 2.65 67 -2.73 -28.3 -3.94 -3.32 68 0.581 -31.8 -3.20 -3.62

127
SUMMARY OF THE RICHARD B. RUSSELL CONCRETE DAM VIBRATION STUDY

Mr. Vincent P. Chiarito

Mr. Robert J. Smith^

ABSTRACT

The forced-vibration tests of the 200-foot high Richard B. Russel gravity dam were conducted

before and after reservoir impoundment. The experimentally measured dynamic properties of

the dam without and with hydrodynamic

interaction were compared to results from three- dimensional (3~D) finite element (FE) modal analyses. Flexible and rigid foundation models

were used in the FE analyses, and the reservoir

was modelled as added mass to the dam. Mode

shapes from the prototype test and FE results

were compared using a modal assurance

criterion. The results reveal the dynamic characteristics of the complex structure with

and without hydrodynamic interaction. The

natural frequencies and mode shapes are consistent with the results of 2-D and 3-D FE analyses. Modal damping ratios were greater

than independently measured concrete specimen

values but were scattered between 1.4 and 5.2 percent. Relative joint motion was observed

above a low response threshold. Ambient

vibrations recorded after impoundment of the reservoir provided another measure of the dam's dynamic characteristics.

KEYWORDS: Ambient Vibrations; Damping Ratios; Finite Element Analysis; Forced-Vibration Test; Frequencies; Mode Shapes

1 . INTRODUCTION

Results of prototype vibration tests (2,3) of

the Richard B. Russell Dam, with and without

reservoir, were compared to linear elastic three-dimensional (3~D) dynamic finite-element

(FE) analyses of the dam. This comparison

provides one basis for assessing the accuracy of

linear elastic FE models for predicting dynamic

properties of concrete gravity dams. The Richard B. Russell Dam, recently built by the Corps of Engineers, is approximately 170 miles (272 kilometers) from the mouth of the Savannah River between Georgia and South

Carolina. A view of the completed dam is shown

in Figure 1. As shown in Figure 2, the crest of

the concrete gravity part of the dam is 1,884 ft long (574 meters). It is composed of 13 nonoverflow, 8 intake, and 11 spillway

monoliths, the tallest of which is approximately

200 ft high (6l meters). The average reservoir

height was 174 ft (52 meters) during the second

test. The grid for the 3-D FE analyses of the dam is shown in Figure 3» A subspace iteration method

was used to solve for the undamped natural

frequencies and mode shapes in the FE analyses

(4). Four FE analyses were conducted: (a) fixed base or foundation, without reservoir, (b) fixed

base, with reservoir, (c) flexible base, without

reservoir, and (d) flexible base, with

reservoir. The flexible base was modelled with

vertical and horizontal springs using an appropriate spring constant. Concentrated

masses were added to model the mass due to the reservoir and the tainter gates. A Modal assurance criterion (1) was used to compare the results of experimental and analytical

structural dynamics. Any attempts to completely

determine the dam-foundation-reservoir

interaction under dynamic conditions are beyond

the scope of this paper.

Richard B. Russel Dam was excited by a 17,000-lb

(7,71 1-kg) inertial mass driven by an electrohydraulic servo-controlled actuator. The

shaker was welded to a 4-ft by 5-ft by 2-in.- thick (1.22 by 1.52 by 0.0508 m) steel plate which was epoxied to the mass concrete of the

dam. The force input to the dam was computed as

the product of the measured acceleration and the 17,000-lb mass. A typical force spectrum of the shaker for the tests is shown in Figure 4-. The maximum stroke of the actuator is limited to 1 in. (0.0254 m). From 25 to 6 Hz (Iz = 1 Hertz =

1 cycle/second), the force spectrum is flat (for

a peak force output of 20 kips (89 kN)). The rolloff from 6 to 1 Hz is due to the stroke

limi tation. Servo-accelerometers with sensitivities ranging

from 0.25 to 5.0 volts/g were used to measure

the horizontal accelerations of the concrete

dam. The accelerometers have a frequency response range from 0 to 500 Hz. Analog signals

were recorded on 32-track magnetic tapes. The

tape recorder has a frequency response range

from 0 to 2,500 Hz. A structural dynamics analyzer was used to later digitize and process

the analog signals. All the recording instrumentation was housed in a closed van positioned on the crest of the monolith adjacent

to the drive point.

U.S. Army Engineer Waterways Experiment

Station, Vicksburg, Mississippi 39180

^ Office, Chief of Engineers, Washington, D.C.

2. EXPERIMENTAL PROCEDURE

128
3. PROTOTYPE TESTS

Two forced vibration tests were made on the Richard B. Russell Dam, before and after

impoundment of the reservoir, to determine the natural frequencies, mode shapes, modal damping

ratios, and relative joint movements. The first low-level forced vibration test was conducted

during January and February 1982 and the second

was conducted during June and July 1984. This

provided an experimental measure of the prototype hydrodynamic interaction and a comparison of the changes of the dynamic

properties.

The dam was excited at monoliths 7, 16, and 22 by a crest-mounted inertial mass (2,3). The

force input was in an upstream-downstream direction. Three arrays of servo accelerometers

measured the horizontal crest accelerations of

31 monoliths in the first test and all 32 monoliths in the second test. Measurements from

drive point 16 overlapped measurements from

drive points 7 and 22. In each test, accelerometers were placed at different

elevations in the three drive point monoliths to measure the horizontal motions in the vertical

planes. The abutments on the sides of the dam restrained

the end monoliths with some degree of fixity. , However, the abutment on the Georgia side was

not complete during the first test and was approximately 60 ft (18. 3m) below the completed

crest elevation. Therefore, a portion of monolith 1 was unrestrained above El. 430.00 on

the Georgia side during the first test.

4. FINITE ELEMENT MODEL

4.

1

Parameter Studies

The grid for the FE model was developed by conducting parameter studies of isolated

nonoverflow, spillway, and intake monoliths

(5). Modal analyses were used to obtain appropriate FE grid sizes for each of the monoliths. A compatible grid was found for all

of the monoliths, thus making possible the generation of the nodal and element data with

the computer. Favorable results were obtained

when modal analyses of each monolith using

uniform grids were compared to analyses using

the grid for the entire 3-D FE model. Figures 5 through 7 show typical sections of the monoliths

and the resulting 3-D grids of each monolith.

In Figure 3 , the resulting 3 -D FE grid of the

dam is shown. Figure 6 b shows the penstock tube opening in the intake monolith modelled as solid, or "smearedhole," elements with 29 percent less stiffness

and density than the mass concrete. Modal

analyses showed good agreement between a finer

FE grid that Included the penstock opening and

the grid shown in Figure 6 b using solid

elements.

4 . 2 Material Properties

The concrete properties used in the FE analyses

were taken from previous measurements on 6- by

12- in. (152- by 305-mm) concrete cylinders.

Dynamic modulus of elasticity values used were

5.1 X 10° psi (35.2 X p° kPa) for exterior mass concrete and 4.26 x 10° psi (29.4 x 10° kPa) for interior mass concrete. From previous foundation tests, the dynamic elastic modulus of the foundation used to compute the spring

constant was 6.05 x 10° psi (41.7 x 10° kPa). Using elastic theory, the average horizontal and vertical spring constant was 4 x 10^^ lb/ ft (5.8 X 10^' N/m) in the flexible base model (6). 4.3 Element Types

Four element types were used to model the dam

(4). Variable-number-nodes (8 to 21 nodes) 3"D thick shell elements were used to model the mass concrete of the nonoverflow. Intake, and spillway monoliths. Plate elements were used to model the concrete spillway piers, the concrete training walls at spillway monoliths 16 and 26, and the concrete deck of the spillway bridge. Beam elements were used to model the steel girders of the spillway bridge. Boundary

(spring) elements were used along the base of

the model in the flexible base analysis.

4 . 4 Boundary Conditions

By definition, the X, Y, and Z translations were defined on the 3~D thick shell elements. All degrees of freedom were deleted along the base of the model for the fixed base analysis. For

the flexible base analysis, the longitudinal (Y) translations and all rotations were fixed along

the base of the model. Except at the dam abutment interface, horizontal and vertical springs were placed at all nodes on the base in the X and Z directions.

In order to model the restraint due to the abutments, nodes along the two end sections

(monoliths 1 and 32) had all degrees of freedom

deleted. However, a few of the nodes had defined translational degrees of freedom on the Georgia side for the model of the dam without

reservoir. As was discussed previously, this abutment was not completed at the time of the first vibration test. For the model without reservoir, a total of 5,127 and 5,609 degrees of freedom were used for the fixed base and flexible base FE models, respectively. For the model with reservoir, a total of 5,100 and 5,582 degrees of freedom were

used for the fixed base and flexible base FE models, respectively.

4.5 Modelling Of Added Mass Concentrated masses were added at appropriate

129
nodes to model the mass due to the reservoir and

the tainter gates. A constant 49.6 ft (15.1m) width mass of reservoir was applied to the model with reservoir. Parameter studies indicated

that this constant width added the same amount

of mass as more refined methods. The mass of

the reservoir acting on the tainter gates and

the mass of the gate was applied at the trunnion

locations on the spillway piers.

5. RESULTS

5.1

Dam Without Reservoir

The 3-D shapes shown in Figure 8 are the first 6 normal mode shapes computed from the FE model

with fixed base. At 7.7 Hz the major response

is associated with the sidesway of the roadway

and concrete spillway pier system. The other

mode shapes show that the major responses of the dam are associated with bending of the dam perpendicular to the longitudinal axis of the

dam. Mode No. 5 at 10.4 Hz is the shape computed by the FE analysis but not measured

experimentally. This was suspected to be due to the possibility of the exciter locations existing at nodes of that mode shape during the vibration test (2). Comparisons of the dam crest mode shapes are shown in Figures 9~12. Note that the mode numbers correspond to bending modes of the dam

crest. The solid symbols represent the normalized mode shape vectors of the experimental results. The solid and dotted

lines, respectively, represent the normalized

mode shapes for the flexible and fixed foundation models. Table 1 summarizes the results of the comparisons in Figures 9-12. Based upon the results of the modal assurance

criterion values, the correlation between

analytical and experimental mode shapes was favorable. The model with a fixed base had slightly better mode shape correlation than the model with a flexible base. For a qualitative

comparison the mode shapes all appear to have

the largest amplitude in the same general location on the crest of the dam. The modal assurance criterion does indicate quantitatively

how good the comparison is among modes. The frequencies estimated by the FE models were higher than frequencies measured

experimentally. This is reasonable since the FE models are an approximation of the dam using a discrete number of degrees of freedom. The FE model in this case is stiffen than the actual

dam. The flexible base model estimated

frequencies closer to the experimentally

measured frequencies than did the fixed base

model. Comparisons of the analytical and experimental

mode shapes in cross section were also

conducted. Results were favorable, with several

modal assurance criterion MAC values greater

than 0.70. The governing mode shape in cross section resembles the fundamental mode shape of

a cantilever beam (8). 5.2 Dam With Reservoir

The 3-D shapes shown in Figure 13 are the first

five major bending modes of the dam crest and

one mode shape associated with sidesway of the roadway and concrete spillway pier system with a fixed base. Note that the mode numbers correspond to the eigenvector number obtained

from analysis. The mode shapes not shown below

8.2 Hz are also associated with sidesway of the spillway pier system. Comparisons of the dam crest mode shapes are shown in Figures 14-1 8. Note that the mode numbers correspond to bending modes of the dam

crest. Table 2 summarizes the results of these comparisons. Mode shape correlation between experiment and analysis was not as favorable as with the dam without reservoir. Both the fixed and flexible base models had approximately the same modal assurance criterion (MAC) values. Various factors may have caused this lower correlation. Anomalies existed in the experimental mode shapes at the lower frequencies (modes 1 and 2) because the force

input was not large enough to completely excite

the entire structure at these frequencies. At the higher frequencies, the experimental data is more reliable because the force input was large enough to excite normal modes. Another source

of error in these experimental mode shapes could

have been the scale factors applied to the measurements from drive points 7 and 22 with respect to the measurements from drive point 16 (3). A least squares method should have been used to obtain scale factors (1). Inadequacies in the FE model assumptions could have produced a lower correlation between experimental and analytical mode shapes. The FE model did not include the effects of the added

mass of the soil at the abutments. The abutment

on the Georgia side of the dam was not complete

at the time the first vibration test was made. The additional soil during the second test might

have caused a slight increase in the stiffness

of the structure at the Georgia-side abutment. Assumptions used for the added mass distribution

of the reservoir may have been slightly

inaccurate, and nonlinearities due to joint

slippage were not modelled. Similar to the case of the dam without

reservoir, the frequencies estimated by the FE models were higher than measured

experimentally. The flexible base model frequencies better compared to the experiment

then did the fixed base model frequencies.

Comparisons of the analytical and experimental

mode shapes in cross section were conducted at

130
two sections. Results were favorable, with

several MAC values greater than 0.90. The governing mode shape in cross section resembles

the fundamental mode shape of a cantilever beam.

It should be noted that some dam crest experimental mode shapes had better correlation

with FE results of a different mode of

vibration. The fourth experimental mode shape

at 7.3 Hz resembled a fourth normal mode of vibration of a fixed at both ends (7). However,

this shape had a better correlation with mode

No. 14 (MAC = 0.63) than with mode No.l6 (MAC = 0.14) from Figure 13. Engineering judgement was

used in interpreting the experimental mode at 7,3 Hz as a fourth mode of vibration. But it is possible that the experimental modes exhibited

measured local behavior not determined by the

model. Also, it is possible there were significant bias errors in the estimates of the mode shape vectors of the spillway and South

Carolina nonoverflow monoliths. In this way,

the MAC can aid in illustrating complex

responses not computed by the model, but measured in the prototype test.

6.

CONCLUSIONS Three-dimensional linear FE analyses have been compared with previous experimental estimates of

the modal parameters of a concrete gravity

dam. Using available dynamic material

properties of the dam concrete and the foundation, a 3~D FE model was successfully

developed to estimate the linear elastic modal

properties of the dam. A mode shape that was

not measured experimentally was computed by the

3“D FE analyses for the dam without reservoir.

The modal assurance criterion was useful for evaluating the correlation between the experimental and the FE mode shapes. Favorable

correlations were computed for the dam without

reservoir, indicating that a reasonable 3-D FE model was developed for computing the first

three natural modes of vibrations. MAC values

ranged from 0.53 to 0.68 for the first three

modes. Less favorable correlations were computed for the dam with reservoir, due to possible errors

in the experimental mode shapes or the FE model boundary conditons. MAC values ranged from 0,15

to 0.44 for the first three modes. Low force'

input at lower fr’equencies caused discontinuities in experimental mode shapes, and

the scale factors applied to the modes may have

been inaccurate. The effects of the soil at the Georgia-side abutment and the reservoir effects

may not have been modelled adequately.

The 3~D FE analyses estimated frequencies higher

than those measured experimentally. This is reasonable since the FE models are an approximation of the dam using a discrete number

of degrees of freedom. The FE model with

flexible foundation predicted natural

frequencies closer to the experimentally

measured values than did the FE model with rigid

foundation.

7. ACKNOWLEDGEMENTS

The results presented herein, unless otherwise

noted, were obtained from research conducted

under the Structural Engineering Research Program of the US Army Engineer Waterways Experiment Station, Vicksburg, Mississippi. Permission was granted by the Office, Chief of Engineers to publish this information.

8. REFERENCES

1. Allemang, R. J., and Brown, D. L., "A Correlation Coefficient for Modal Vector

Analysis," Proceedings of the 1st International Modal Analysis Conference,

1982, pp 110-116.

2. Chiarito, Vincent P., and Mlakar,

Paul F. "Vibration Test of Richard B. Russell Concrete Dam Before Reservoir

Impoundment," TR-SL-83-2, May, 1983.

3. Chiarito, Vincent P., Recent results of

the second vibration test of Richard B. Russell Concrete Dam after reservoir

Impoundment (Final report in preparation).

4. Bathe, K. J., Wilson E. L., and Peterson,

F. E., "SAPIV: A Structural Analysis Program for Static and Dynamic Response of Linear Systems," University of California, Earthquake Engineering Research Center, Richmond, California, 1973.

5. Chiarito, Vincent P. "Linear Finite Element Comparison with Experimental Modal Analysis for a Concrete Gravity Dam," Proceedings of the Third International Modal Analysis Conference , Vol. 1, 1985, pp 59-65.

6. Timoshenko, S. P., and Goodier, J. N., Theory of Elasticity , 3rd ed., McGraw-Hill,

New York, 1970.

7. Shock and Vibration Handbook , C. N. Harris, and C. E. Crede, eds, McGraw-Hill, New York,

1961 . 8. R. Stephen Wright and Vincent P. Chiarito, "Comparison of Experimental and Analytical Structural Dynamics Using a Modal Assurance

Criterion," from SAE SP-635-Structural

Dynamic Testing and Analysis , SAE Technical

Paper Series 851928, Presented at Aerospace Technology Conference and Exposition, Long

Beach, California, Oct 14-17, 1985.

131
Table 1 . Crest Mode Shape Comparisons - Dam without Reservoir

Prototype 3-D FE Analyses

Crest V ibration Fixed Flexible

Mode Test Base Base Number Frequency (Hz) Frequency (Hz) MAC* Frequency (Hz) MAC*

1 5.9 7.1 0.67 6.6 0.68

2 6.8 7.9 0.55 7.4 0.53

3 7.6 9.3 0.58 8.7 0.55

5 8.2 10.7 0.43 10.0 0.13

* Modal Assurance Criterion (1)

Table 2. Crest Mode Shape Comparisons -Dam with Reservoir

Prototype 3-D FE Analyses

Crest Vibration Fixed Flexible

Mode Test Base Base

Number Frequency (Hz) Frequency (Hz) MAC* Frequency (Hz) MAC*

1 5.3 6.2 0.20 5.7 0.19

2 6.2 6.8 0.44 6.4 0.38

3 6.6 7.8 0.17 7.3 0.15

4 7.3 8.2 0.14 7.6 0.23

5 8.1 8.7 0.09 8.0 0.33

* Modal Assurance Criterion (1)

Figure 1 . View of the Dam from the Downstream Georgia Side

132
133

Figure

2.

Plan

and

Elevation

Views

of

Richard

B.

Russell

Dam
MONOLITHS
1 2 4 6 8 10 20 FREQUENCY, Hz Figure 4. Typical force spectrum - force output of electrohydraulic

vibrator versus frequency

€ SERVICE ROAD AND AXIS OF DAM

TYPICAL NONOVERFLOW model

a. Cross-section

Figure 5. Typical Nonoverflow Monolith

135
TYPICAL INTAKE

a. Cross-section

PENSTOCK TUBE OPENING (SHADED AREA)

bo Finite-element model

Figure 6. Typical Intake Monolith

I SERVICE ROAD AND

^ AXIS OF DAM

Figure 7. Typical Spillway Monolith

136
^ ^ ' r: ; MODE NUMBER 1, 7.1 Hz

c ^ r rrrj^-33

MODE NUMBER 2, 7.7 Hz

MODE NUMBER 3, 7.9 Hz

MODE NUMBER 4, 9.3 Hz

T \ 1 JJJU'^ N 1 1 1 1 ^

1

- ir ^ MODE NUMBER 5, 10.4 Hz

Figure 8. FE Model Mode Shapes - Dam Without Reservoir

137
MONOLITHS MONOLITHS

Figure 9. Mode 1 - Dam without Reservoir Figure 10. Mode 2 - Dam without Reservoir

Figure Mode 3 - Dam without Reservoir Figure 12. Mode 5 Dam without Reservoir

138
Pl^FF ifIt

L-l-MODE NUMBER 1, 6.2 Hz

MODE NUMBER 2, 6.3 Hz

MODE NUMBER 11, 6.8 Hz

MODE NUMBER 14, 7.8 Hz

MODE NUMBER 16, 8.2 Hz

1

1—

L=(= MODE NUMBER 17, 8.7 Hz

Figure 13. FE Model Mode Shapes - Dam with Reservoir

139
Figure 14. Mode 1 - Dam with Reservoir “1

legv~END 0--0

EXPEmMENTAL

PE. FIXED BASE

F E. flexible base

1 n \

.

\ •

\

1 • \

,

,JUUl \‘ fl u

\ . /

m' /

r \

n / \

;

r

i

“ /

d0,30 0.40 0.00 OJO NOf^MAUZCO LfNQTH

0 OJO 0.40 0.60 OJO 1.00 MORMAU2CO LENGTH Figure 16. Mode 3 - Daa with Reservoir Figure 17. Mode 4 - Dam with Reservoir

Figure 18. Mode 5

- Dam with Reservoir

140
SHAKING TABLE TESTS ON SEISMIC STABILITY OF

REINFORCED EMBANKMENTS

BY

1 2 Yasuyuki Koga , Eiichi Taniguchi

and Yoshihiro Ito^

ABSTRACT

Recently geotextile has been used for reinforcing embankments, but there remain some unknown

points on the aseismicity of geotextile reinforced embankments. Shaking table tests were performed to clarify seismic effects on embankments reinforced by non-woven fabric. The

tests consisted of two cases; reinforced em- bankments on liquefiable sand deposits and

those on inclined ground. Test results showed

that the settlement of a reinforced embankment

due to cyclic loadings was smaller than that of

a non-reinforced one. Moreover the reinforced

embankment settled as a unified soil block during excitation and therefore it kept its original shape after it settled and very few cracks

were observed on the surface of the embankment.

KEYWORDS: Earthquakes; Embankments; Geotextiles

1. INTRODUCTION Recently geotextile has been used for reinforcing embankments, but there remain some unknown

points on the aseismicity of geotextile embankments reinforced . Shaking table tests were performed to clarify seismic effects on embankments reinforced by non-woven fabric. The

tests consisted of two cases; reinforced em- bankments on liquefiable sand deposits and

those on inclined ground. Measurements carried

out during the tests included acceleration,

pore water pressure, displacement of the ground

and embankment, and tension of the non-woven

fabric. Test data were compared in reinforced

and non reinforced cases.

2. PURPOSE

The purposes of shaking table tests described

in this paper are:

(1) to determine if the reinforcement by non- woven fabric to embankments is effective

against earthquakes,

(2) to understand the vibratory behavior and final failure pattern of embankments reinforced by a non-woven fabric during cyclic

loadings.

3. PROCEDURE

3 . 1 Outline of Models

Fig. 1 shows the cross sections of four models

used in the shaking table tests. Among these

four models Models L-1 and L-2 have embankments

on a liquefiable saturated sand deposit, and Models S-1 and S-2 have embankments on an in- clined stiff ground.

The embankments and ground in Models L-1 and L-2 are constructed by using Sengenyama sand. Table 1 shows the physical properties of Sengenyama sand. Models L-1 and L-2 were pre- pared as follows: air dried Sengenyama sand was spread in a steel box at the thickness of 20 - 30 cm for ground and 10 cm for embankment,

and compacted by men’s feet, and then water was infiltrated from the bottom of the ground to saturate it. The ground of Model L-1 is in a loose condition with relative density of 53 - 62 percent, and the ground just beneath the em- bankment of Model L-2 is compacted to a dense condition with relative density of about

80 percent. The left side embankment is reinforced by placing non-woven fabric at 10 cm intervals; the right side is not reinforced. Table 2 shows the physical properties of the non-woven fabric.

The ground of Models S-1 and S-2 was prepared

by fully compacting sandy silt (Table 1) by a wooden block to eliminate destruction by excitation. The embankment was made by 10 cm of Sengenyama sand that was compacted by men's

feet. Its relative density is 43 - 65 percent. Both Models S-1 and S-2 have a reinforced em- bankment on the left side and a non-reinforced

one on the right side. Although the boundary

surface between the embankment and the ground

is smooth in Model S-1, the boundary of Model S-2 is of bench shape. The non-woven fabric

is fixed by 15 cm long nails at 5 cm intervals.

3.2 Scaling Law

The following scaling law was adopted in this series of tests; the reinforcement ratio R (non dimensional) ( 1 ) is defined by the following equation. It is the same for the prototype

and the model.

^ Head

2 Senior Research Engineer

^ Research Engineer, Soil Dynamics Division,

Public Works Research Institute, Ministry of Construction, Tsukuba Science City, Ibaraki

305 Japan

141

\
where average tensile strain of reinforced soil in horizontal direction < 0) E : Young's modulus of reinforcing material

t : thickness of reinforcing material

horizontal confining pressure

acting on reinforced soil

aH: placement interval of reinforcing material

We assume that the reduced scale of the model

is "7 and of the model agrees with that of the prototype. Thus, the following equations

can be drawn.

7<»3oV •

/ E.t \

(">m \, ^30 • /

E.t \ 030 • AH

1 (E.t)p 49 ( 2 )

The Young's modulus and the thickness of the non-woven fabric used in the prototype produced

values as in Table 3; therefore Eq. (2) holds

for non-woven fabric used in the tests.

3 . 3 Shaking Condition

Table 4 shows the shaking condition. The di- rection of shaking is only horizontal and lon- gitudinal to the model. Tests No. L-1-1 - No. L-1-3 were performed by using one model.

Test No. L-1-1 was performed and after pore water pressure in the model was dissipated,

test No. L-1-2 was conducted, and then test

No. L-1-3 was performed. Other tests were carried out in a similar way.

3 . 4 Measurement

In models L-1 and L-2, acceleration, pore water

pressure, settlement of the ground surface, and tension of the non-woven fabric were measured

at 16, 35, 3 and 3 points, respectively. In models S-1 and S-2, acceleration, settlement of the ground surface, tension of the non-woven

fabric, and displacement of the non-woven fabric were measured at 16, 3, 3, and 3 points

respectively. Small targets were placed on the surfaces of all models and the vertical and horizontal displacements were measured before

and after each shaking test. The deformation

of the model was observed through an acryl side

panel; photographs were taken during

excitation.

4. RESULTS AND DISCUSSIONS

4 . 1 Reinforced Embankment on Liquefiable Sand Deposits

Test results of Models L-1 and L-2 are de- scribed in this section. Fig. 2 shows the re- lationship between the acceleration on a shaking table and the cumulative settlement of the embankment. The settlement of an embankment

reinforced by non-woven fabric is smaller than

that of the non-reinforced one as illustrated

by Models L-1 and L-2. In the test case for a shaking table acceleration of 173 gal, the set- tlement of reinforced embankment of Model L-1

is about half of that of non reinforced one. The settlement of reinforced embankment on the improved ground of Model L-2 is about one fifth

of that of non-reinforced embankment on non- improved ground shown in Model L-1. These test results indicate that reinforcement by non- woven fabric is effective in reducing the set- tlement of embankments caused by liquefaction

of the subsoil.

Fig. 3 shows the deformation of the side mesh

of models in cases of Tests No. L-1-3 and

No. L-2-4. The figure denotes that the defor- mation of the ground in Test No. L-1-3 occurs

in the area above the line 30 cm high from the bottom. Ground deformation from Test No. L-2-4

occurs in the area above the line 70 cm high

from the bottom resulting from ground improvement under the embankment. Table 5 shows average horizontal strain at positions of A - J shown in Fig. 3. It was calculated from the displacements of intersections of the mesh before and after excitation. Table 5 indicates that the average horizontal

strain in the reinforced embankment and the shallower part of its subsoil in Test No. L-1-3

is reduced to 26 - 42 percent of that in the non-reinforced one. In Test No. L-2-4 the av- erage horizontal strain is decreased to 14 -

61 percent. It is shown that non-woven fabric, horizontally spread in the embankment, is able

to reduce the horizontal tensile strain caused

by cyclic loadings in the embankment and subsoil; this results in the reduction of the settlement of the embankment. Photographs 1 and 2 show the view of the top of reinforced and non-reinforced embankment, re- spectively, after shaking in Test No. L-2-4.

The non-reinforced embankment is considerably

cracked and completely destroyed. On the other hand the embankment reinforced by non-woven

fabric has few cracks and keeps its original

shape; the effect of fabric reinforcement is clearly observed. It is shown from the test

that an embankment reinforced by non-woven fabric settles if the subsoil is liquefied, but keeps its original shape because the embankment

is unified by the non-woven fabric.

142
Figs. 4 and 5 show the change of acceleration,

pore water pressure, settlement of embankment

and tension of non-woven fabric with time in Tests No. L-1-3 and No. L-2-3. Fig. 6 shows

the position of sensors and the ratio of pore water pressure to effective overburden pressure

10 seconds after the beginning of excitation.

In Fig. 4 the accelerations at A13 and A14 have

almost the same value before the ground begins

to liquefy, 2 seconds after the beginning of excitation, and therefore there are almost no effects of reinforcement on the acceleration in

the embankment. As the pore water pressure at

P7 and P13 starts to increase 2 seconds after

the beginning of excitation, the settlement of embankment D1 and D3 appears. The value of pore water pressure at P7 is about half of that

at P13. A reason for it is that shear strain

is not likely to increase in the ground just under the embankment because of confining

effects by non-woven fabric. In Fig. 4 the tension T1 at the boundary between embankment

and ground is four times larger than the ten- sion T2 and T3 at the upper part of embankment.

This result implies that if an embankment on liquefiable ground is reinforced by geotextile,

it is necessary to decrease the placement in- terval of geotextile or place stronger geotextile in the lower part of the embankment. The value of tension T1 increases with time, but

the value of tension T2 starts to decrease when

the embankment begins to settle 2 seconds after

the start of excitation; it then begins to in- crease again.

Fig. 5 shows that the pore water pressure P7 in the ground just under the reinforced embankment

starts to increase 9 seconds after the beginning of excitation and reaches the peak value

at 15 seconds, but that the pore water pressure

P13 just under the non-reinf orced embankment

starts to increase 4 seconds after the beginning of excitation and reaches the peak value

at 9 seconds. As the pore water pressures P7 and P13 increase, the reduction of acceleration

A13 and A14 and the settlement of embankment D1 and D3 appears. The settlement D3 at the end

of excitation, 8.8 cm, is twice as large as

that of Dl, 4.2 cm. Thus, the generation of pore water pressure just under the reinforced

embankment is restrained and the required time

to reach initial liquefaction becomes longer. This phenomenon is presumably due to the deformation of shallower ground being restrained by

the fabric reinforcement. However such a phe- nomenon is not clearly observed in Model L-1, and so further investigation is needed. The tensions T2 and T3 increase with time until

15 seconds after the beginning of excitation

when the embankment settlement suddenly in- creases and then the value T2 increases and T3 decreases

.

Fig. 6 shows that the ratio of pore water pressure to effective overburden pressure u/o' in the ground under the reinforced embankment is smaller than that under the non-reinforced one

in Tests No. L-1-3 and No. L-2-3. In Test

No. L-2-3 the value of u/a' in the improved

area is smaller than in the non-improved area.

Figs. 7 and 8 show the change of the amplification ratio of acceleration with time. It can

be noted from Figs. 7 and 8 that the amplification ratio of the reinforced embankment takes almost the same value as the non-reinforced one until the ground is liquefied. In Fig. 7 the acceleration amplification ratio of the reinforced embankment is larger than that of the non-reinforced one at 10 seconds after the be- ginning of excitation. This may caused by the effect of the inclination of the accelerometer

in the liquefied ground. In Fig. 8, the delay

of time when the initial liquefaction occurs in the ground under the embankment relates to the delay of time when the acceleration begins to decrease

.

4 . 2 Reinforced Embankment on Inclined Ground

This section describes the test results of Models S-1 and S-2. Fig. 9 shows the deformation

of the embankment after excitation with accel- eration of about 850 gal. The result of Test

No. S-1-6 indicates that almost the same value

of settlement is induced by cyclic loadings to both reinforced and non-reinforced embankments. Actually the top of the slope of non-reinforced

embankment is heavily destroyed, but the rein- forced embankment is unified by the non-woven

fabric as it settles as a single block of soil. Same failure pattern is seen in Test No. S-2-7. A considerably smaller number of cracks were observed on top of the reinforced embankment

than on the top of non-reinforced one. It is observed from the comparison of the failure pattern of non-reinforced embankment in Test

No. S-1-6 and No. S-2-7 that a large failure occurred in both cases at the top of the slope. However in Test No. S-1-6 a slip surface exists between the embankment and the inclined ground.

A large slip occurs there, but in Test No. S-2-7 such a large slip is not observed because benching occurred between the embankment and the in- clined ground. This difference appears more clearly for embankment reinforced by non-woven

fabric; although in Test No. S-1-6 a clear slip surface is seen between the embankment and in- clined ground. No clear slip surface is ob- served in Test No. S-2-7. The embankment holds

its original shape because non-woven fabric is nailed on benches. A small slip is observed between the soil and the non-woven fabric. A smaller slip is observed in Model S-2 over S-1 where non-woven fabric is fixed on benches. Moreover if non-woven fabric is rolled into the embankment in the front of the slope, the slip will be reduced.

Fig. 10 shows the relationship between the ac- celeration on shaking table and the settlement

of the embankment. The settlement was calculated by averaging the settlement of targets on

the top of the embankment. Any type of embankment in Fig. 10 greatly settles when the accel- eration reaches about 850 gal. A comparison

143
of the settlement of embankment in the case of acceleration of about 850 gal indicates that both the reinforced and non reinforced embankment have almost the same average settlement in Model S-1, but that the average settlement of

the reinforced embankment in Model S-2 is con- siderably smaller than that of the non- reinforced one.

Fig. 11 shows the displacement of non-woven

fabric during cyclic loadings in Test

No. S-1-5. This figure shows that the non- woven fabric in the upper part of embankment

first starts to move and every non-woven fabric

does not start to move at the same time.

Fig. 12 shows the relationship between the ac- celeration on the shaking table and the maximum

tension of non-woven fabric. It indicates that

the tension increases with the increase of ac- celeration. The lower the position of non- woven fabric in the embankment, the larger tension occurs in Model S-1 in the case with

acceleration of about 850 gal in which large deformation appears. In Model S-2 the maximum

tension occurs at the position of No. 1 of the non-woven fabric which is located in the lowest

part of embankment, and the second largest

tension occurs at the position No. 5 of the non-woven fabric which is located in the highest

part of the embankment. This is due to the effect of fixing the end of non-woven fabric on benches.

5.

CONCLUSIONS

B. Reinforced embankment on inclined ground:

(1) An embankment on inclined ground is unified by reinforcing by non-woven fabric

and the reinforcement can prevent the failure at the top of slope.

(2) The settlement of the embankment due to cyclic loadings can be considerably re- duced by benching the inclined ground and fixing non-woven fabric on the benches.

(3) The maximum tension occurs at non-woven

fabric in the lower part of the embankment

.

6. ACKNOWLEDGEMENT

The authors express their appreciation to Mr. T. Morishita, Mr. T. Shimazu, Mr. S. Washida for their help during the shaking table

tests

.

7. REFERENCES

(1)

Tatsuoka, F. et al.: Reinforcement of clayey embankment by non-woven fabric, Tsuchi-To-Kiso, Vol. 33, No. 5, 1985 (in Japanese)

.

A. Reinforced embankments on liquefiable sand deposit

:

(1) Embankments reinforced by non-woven fabric settle to some extent when the subsoil is liquefied. However the settlement of the reinforced embankment was about half that

of a non-reinforced one. The settlement

of reinforced embankment on the improved

ground was about one fifth of a non- reinforced embankment on non-improved

ground. It is found that the embankment

reinforced by non-woven fabric is unified

and deforms as a single soil block and the embankment is hardly cracked by cyclic

loadings

.

(2) Non-woven fabric is able to restrain the horizontal tensile strain in embankment

and ground.

(3) The pore water pressure in the ground under the reinforced embankment is smaller

than that in the ground under the nonreinf orced one.

(A) The tension acting on the non-woven fabric

in the lower part of embankment is about

four times as large as that in the upper

part of the embankment.

144
Table 1 Physical properties of material

soil name Sengenyama sand sandy silt

Japanese unified soil classification SPu ML

specific gravity Gs

1 2.71 2.70

average grain size Dso (mm)

1 0.28 0.057

uniformity coefficient Uc 2.91 39.6

fine content rate (< 74 jum) (%) 2.4 53.0

Table 2 Properties of non-woven fabric

material

1 nylon 70 %

polyester 30 %

thickness (mm) 0.217

tensile strength at peak Of (kgf/cm^) 13.2- 13.6

strain at peak (%) 38-29

Young’s modulus at e=15% Eis (kgf/cm^) 47.3 - 57.3

Table 3 Data for non-woven fabric of model and prototype

model prototype (Et)^

Young’s modulus E (kgf/cm^)

j

thickness

t (mm)

Young’s modulus E (kgf/cm^)

thickness

t (mm)

(E %

56.3 i 0.217

i

200 1 3.08

1 50.4

150 4.49

1 55.

1

145
Table 4 Shaking condition

test model

No. No.

acceleration on

i the shaking table

(gal) frequency

(Hz)

duration time

(sec) L-1-1 25 L-1-2 L-1 j 113 L-1-3 1 173 L-2-1 i 25 2 1 20 L-2-2 ; 117 1 I L-2-3 ! L-2 173

i L-2-4 1 240 ii S-1-1 26 1

i S-1-2 123

1 S-1-3 248 S-l-4 S-1 395

S — 1-.5 580 S-1-6 '

! 850 S-2-1 !

' 4 10 S-2-2 ! 122 S-2-3 i

! 245 S-2-4 S-2 400 S-2-5 ] 590 S-2-6 1

1 '00 S-2-7 840

Table 5 Average horizontal strain

test No.

average horizontal strain (%)

reinforced 1 non 1reinforced L-1-3 A ! -2.4 F -4.0 B 1.6 G 3.8 C 3.2 H 12.3 D 1 11.4 ! 1 8.2

i E i 0.9 1 J

i 0 L-2-4 A 4.0 F i 28.0

1 B 18.9 G 33.0 C 13.2 H 1 21.8 D 0 1 1 0 E 0 1 _JJ 0

note : e[, positive : tension

negative : compression
model

]

]

]

1

_l

1

]

]

]

]

]

]

]

]

]

]

1

1

I

L-2

52.1

non-woven

fabric

o 78.3 79.3 sand

54.1 78.9 j 50.4 -—compacted —

•

—— compacted —

1.3 0.75 1 0.5

1 0.75 1.4 0.75 1 0.5 0.75 1.3

Figures in the models show relative density in per cent.

Fig. 1 Models used in tests

acceleration on shaking table! gal)

Fig. 2 Cumulative settlement of embankment and acceleration on shaking table

147
A13 gal) acceleration

A14 ' gal

'

acceleration

P7 ' gf . dll'

)

pore water

pressure P13 (gf/ciD') pore water

pressure D1 cm) settlement of embankment

D3 cm) settlement of embankment

T1 kgf)

tension of fabric T2 kgf)

tension of fabric

T3 kgf.

tension of fabric

Fig. 4

Test No.L-1-3

A. F _

B.G

^ 'rr'

L_.

L

^

1 L F 11

i

' B 1 G 1 C.D.E H.I.J before

excitation

after excitation (A~J show the place where average horizontal strain Ch was calculated.!

Fig. 3 Failure pattern of models

0 10 : time sec Records of measurements (Test No. L-l-3)

400 A13 (gal) acceleration -400

400 A14

acceleration 0 -400

P7 (gf cm*) 200 pore water

pressure 0 P13 (gf uir) 200 pore water

pressure

0 D1 .cm) 0 settlement of embankment 5 D3 ( cm )

1 settlement of embankment 10 ‘

T2 kgf)

tension of

fabric

0

‘

T3 kgf

2 tension of fabric

0

Fig. 5 Re

10 time sec Hg. 5 Rerords of measurements (Test No. L-2-3)

148
]

]

]

]

]

]

]

]

]

]

]

]

]

]

]

Fig. 6 Distribution of rate of pore water pressure to effective overburden pressure

reinforced non reinforced

amplification ratio of acceleration Fig. 7 Amplification ratio of acceleration (Test No. L-1-3)

reinforced non reinforced

amplification ratio of acceleration Fig. 8 Amplification ratio of acceleration (Test No. L-2-.3)

a ) Test No.S — 1 - 6

b ) Test No.S — 2 - 7

after before —• slip surface

excitation excitation

Fig. 9 Deformation of embankment after excitation

149
cumulative

settlement

of

embankment

acceleration on shaking table gal)

0 200 400 600 800 1000

Fig. 10 Cumulative settlement of embankment

and acceleration on shaking table

' 1 ' 1 ) 1 1 1 ' 1 ^ time' sec'

Fig. 11 Displacement of non-woven fabric during excitation

Fig. 12 Maximum tension in non-woven fabric and acceleration on shaking table

Photo. 1 View of top of reinforced embankment

after excitation (Test No. L-2-4) Photo. 2 View of top of non reinforced embankment

after excitation (Test No. L-2-4)

150
STATUS REPORT ON NEW STUDY OF LOWER VAN NORMAN DAM

BY

A. G. Franklin^

ABSTRACT

This study consists of a re-analysis of the slide that occurred in the Lower Van Norman Dam

during the 1971 San Fernando earthquake. In addition to information obtained in earlier in- vestigations, new data are being obtained from

field and laboratory investigations of the hy- draulic fill sand in the remnant of the original dam. The objective is to validate a method

of evaluating the post-earthquake stability of dams with liquefaction-susceptible soils, using

either steady-state strength or some other form

of residual strength. The approach is to compare strength values from field and laboratory

tests of materials in the remnant of the original dam with those that can be inferred from

its observed behavior during and immediately

after the 1971 earthquake.

KEYWORDS: Earthquakes; Dams; Soils; Liquefaction

1. INTRODUCTION

The Lower Van Norman Dam, also known as the Lower San Fernando Dam, was severely shaken by

the San Fernando Earthquake of 9 February 1971, and suffered a major slide of the upstream

shell and crest. These events and the characteristics of the dam were studied exhaustively

and reported by Seed, et al. (1975a, b). The investigations showed that the slide occurred

as a consequence of earthquake-induced liquefaction of a zone of hydraulic fill sand in the upstream shell.

The dam was of hydraulic fill construction in

its lower portions, with about 45 ft (14 m) of rolled fill at the top (Fig. 1). The slide which followed the 1971 earthquake resulted in

the loss of a major part of the crest and the upstream shell (Fig. 2), and loss of the reservoir was averted only because the water level

was fortuitously at a relatively low level.

The dam was subsequently rebuilt, incorporating

the remnant downstream of the slide. A replacement dam was built upstream, so that the original dam now serves only as an emergency

detention structure.

The Corps of Engineers, with Professor H. Bolton Seed and Geotechnical Engineers,

Inc. (GEI) , is presently conducting a new study

of the dam and the 1971 slide, using the infor- mation provided by the original investigations

and new data being obtained from field and lab- oratory investigations of the hydraulic fill

sand in the still extant remnant of the original dam.

2. PURPOSE AND OBJECTIVES Conventional methods of evaluating liquefaction

potential in embankment dams deal primarily

with the development of excess pore water pressures in embankment and foundation soils, to predict whether or not excess pore pressure buildup and loss of shear strength will occur

in the event that the structure is shaken by an earthquake of a size judged to be possible at the site. If the outcome of the analysis is the conclusion that such pore pressure buildup will occur, it is frequently assumed that the structure is therefore unsafe. However, very

large strains can occur only if the shear strengths are reduced to values below the shear stresses produced by the static loading conditions. A complete evaluation of seismic sta- bility thus requires the determination of the minimum values of shear strength that could be developed under conditions of liquefaction and large strain.

The earlier investigations of the 1971 slide

provided, from field observations, particularly

well-defined and well-documented description of the failure mechanism, including the zone of soil liquefaction and the locations of the slip surfaces. These observations provide a basis

for obtaining from a stability analysis an upper bound value for the undrained residual

shear strength of the soil in the liquefied

zone, and an estimate of the shear strength at the time the slide was initiated. In the present investigation, the properties of the hy- draulic sand fill are being re-examined to determine whether field and/or laboratory tests can be used to predict the post-liquefaction or residual strength, and so to determine the ul- timate stability of the structure.

One of the participants in the current study is Geotechnical Engineers, Inc., whose Principals Gonzalo Castro and Steve Poulos have developed

a theory and analytical procedure based on the concept of the steady state of deformation

(Poulos, 1981). It is considered chat "lique- faction" is Che onset of continuing, large de- formations under shear stresses that exceed the available shear strength (Castro, 1975). The

Chief, Earthquake Engineering and Geophysics

Division, Geotechnical Laboratory, U.S. Army Engineer Waterways Experiment Station, Vicksburg, Mississippi

151
undrained steady-state strength is defined as the shear stress in a saturated soil as it undergoes continuing deformation at constant

void ratio and strain rate; and it is achieved

only after liquefaction has occurred. The steady-state strength is similar to a residual undrained strength. It is believed to be de- pendent only on the void ratio of the soil, and

is defined by the steady-state line on a plot

of void ratio vs. initial effective minor principal stress, or approximately, the undrained

steady-state strength (Fig. 3); obtained from strain-controlled, undrained, monotonically

loaded shear tests. Because the steady-state

strength is very sensitive to void ratio, accurate determination of the in-situ void ratio or density is essential in applying this theory. Detailed discussions of the application of the theory to liquefaction evaluation and to the lower Van Norman Dam are given by Poulos (1985) and Castro (1985)

.

Seed (1985, unpublished) has used back- calculation of shear strengths from several

past occurrences of liquefaction to obtain a tentative correlation of residual strength with

SPT "N" values (Fig. 4). This offers the pos- sibility of using in-situ tests for evaluation

of post-liquefaction strength. The Cone Pene- trometer Test (CPT) may also prove useful for

the purpose, either through correlation with

the SPT or by direct comparison of CPT test values with back-calculated residual strengths.

The objective of the present study is to validate a procedure for evaluation of post- earthquake stability of embankment structures

using post-liquefaction shear strengths ob- tained from field or laboratory tests. The approach being used is to compare the strength

va..aes obtained in the remnant of the original

Lower Van Norman Dam with those that can be in- ferred from its observed behavior during and immediately after the earthquake of 9 February

1971.

3. PLAN AND PROGRESS OF STUDY

3 . 1 Review of Historical Data

Prior to initiation of field work, a review of historical records, including settlement and hydrological data, was done by GEI at the offices of the owner, the Los Angeles Department of Water and Power, and at State Government offices in Sacramento, California.

3 . 2 Field Investigations

The field phase of the study was completed in December 1985. Twelve CPT soundings were performed in September 1985, to depths ranging

from 83 ft (25 m) to 123 ft (37 m) through the core and the downstream shell, but with most located to test shell materials. Six borings

for split-spoon samples and SPT's were done by

a Waterways Experiment Station (WES) drilling

rig, and the SPT system was later calibrated

using dynamic energy measurements. Six additional borings were made to obtain high-quality undisturbed samples, using a Hvorslev fixed- piston sampler. Two more borings were drilled

as observation wells. A test shaft of 6-ft diameter (2 m) diameter

was excavated through the downstream shell.

(The location is shown in cross-section in Fig. 2.) This shaft provided access for visual

and photographic examination of the shell materials, to take high-quality undisturbed samples

by hand carving into tubes, to do in-place

density tests, and to take remolded bag samples

.

X-radiographs were made of all of the undisturbed samples, to evaluate any possible ex- cessive disturbance and as an aid in selecting

sample intervals for testing.

3.3 Laboratory Testing

Laboratory testing is being done at the laboratories of GEI, in Winchester, Massachusetts; at Stanford University; and at WES. Laboratory

tests will include the following:

Atterburg limits; Grain-size distribution;

Compaction; Maximum and minimum density;

Cyclic loading triaxial tests; Monotonic loading triaxial tests. The cyclic loading and monotonic loading tri- axial tests will be done on both undisturbed

and remolded specimens. Laboratory testing is scheduled to be completed by the end of September 1986.

3 . 4 Analysis and Report

The final phase of the study is to be the analysis of the data and of the failure mechanism

during the slide, from October 1986 to January

1987, and preparation of the final report, scheduled for completion in July 1987.

4. REFERENCES

(1) Castro, Gonzalo; Poulos, Steve J.; and Leathers, Francis D. 1985. "Re- Examination of the Slide of Lower

San Fernando Dam," Journal of Geotechnical

Engineering , Vol. Ill, No. 9, pp. 1093-

1107.

(2) Castro, Gonzalo 1975. "Liquefaction and Cyclic Mobility of Saturated Sands," Jour - nal , ASCE, Vol. 101, No. GT6, pp . 551-569.

(3) Poulos, Steve J. 1981. "The Steady State

of Deformation," Journal , ASCE, Vol. 107, No. GT5, pp. 553-562.

152
(4) Poulos, Steve J. 1985. "Liquefaction

Evaluation Procedure," Journal of Geotechnical Engineering , ASCE, Vol. Ill,

No. 6, pp. 772-792.

(5) Seed, H. B.; Lee, K. L.; Idriss, I. M.

;

and Makdisi, F. I. 1975a. "The Slides of the San Fernando Dam During the Earthquake

of 9 February 1971," Journal , ASCE,

Vol. 101, No. GT7, pp. 651-689.

(6) 1975b. "Dynamic Analyses of

the Slide in the San Fernando Dam During

the Earthquake of 9 February 1971," Jour- nal , ASCE, Vol. 101, No, GT9, pp. 889-911.

Figure 1. Cross-section through Lower Van Norman Dam

(Los Angeles Department of Water and Power)

Figure 2, Cross-section through Lower Van Norman Dam, showing location of base of 1971 slide and location of test shaft.

153
VOID

RATIO,

T

0.« •

0.4 -

CONTRACTIVE STATE

0.2' ' S —

10 100 1000 10,000

UNDRAINEO STEADY STATE SHEAR STRENQTH, p«f

Figure 3. The Steady-State Line

^^1^60

Figure 4. Tentative Relationship between Residual Strength

and SPT N values. (After Seed, 1985, unpublished)

154
SEISMIC DESIGN PROCEDURE AND DYNAMIC STRENGTH OF CAB

BY Toshio Iwasaki, Kazuhiko Kawashima, Naomi Obinata, and Shigeki Unjoh

ABSTRACT

The installation of CAB (Cable Box) is initiated to enhance underground distributing

network services of electricity and telephone

at populated cities in Japan. Draft Manuals

for Design of CAB Systems were produced by the Ministry of Construction. This paper briefly

describes the seismic design procedures and the dynamic load test results performed to verify

the seismic performance of a standard CAB

element

.

1 . INTRODUCTION

Special reinforced concrete ducts called CAB (abbreviation of cable box) housing trans- mitting lines (network service of electricity

and telephone) are being constructed under city

roads in Japan. The CAB network systems is generally more expensive than the current

transmission systems which hangs the wires from

electric poles. However, CAB has advantages

such as disaster prevention and an increases

available road space for service (refer to Fig.

1 ). CAB has a standard cross section (see Fig. 2 ). One element of CAB is composed of an U-shaped

duct with length of approximately 1 . 5 m; it has

a concrete cover. Photographs 1 and 2 show an installation of the U-shaped duct and the

cover, respectively. The Ministry of Construction developed the design procedures including

the "Manuals for Design of CAB Systems

(Draft)." In conjunction with examining

seismic design procedures, a series of dynamic

loading tests were performed to verify the seismic performance of CAB systems at the Public Works Research Institute. These tests were at the request of the Road Bureau and the Kantoh Regional Construction Bureau of the Ministry of Construction.

This paper presents a stjumnary of the proposed

draft seismic design procedures in the Manuals

for Design of CAB Systems and the dynamic

loading test results.

2 . OUTLINE OF SEISMIC RESISTANT DESIGN

PROCEDURE

The Manuals for Design of CAB Systems (Draft), hereafter referred as the manuals, stipulate

that the design of CAB must resist seismic

effects. CAB may be installed at soft ground

Earthquake Disaster Prevention Department.

sites and/or sites vulnerable to liquefaction.

Seismic design testing is required for:

1) deformation of surrounding ground

developed during earthquake and excessive

pore water pressure,

2 ) longitudinal direction of CAB axis,

3 ) uplift when the surrounding subsoils are vulnerable to liquefaction.

2 .

1

Seismic Ground Strains Considered in Seismic Design

Designing for deformation of subsurface ground

during earthquakes is accounted for in terms of seismic ground strains «q, and is represented

as €g - i'l X 1/2 X 1/3 X Eq (1) in which 1/^ , and 1/3 represent the seismic

zone factor, soil condition factor and lique- faction effect factor, respectively, and Cg represents a standard seismic ground strain

(-1/ 100 ) . Factors and 1/2 are specified in Tables 1 and 2 , respectively, in accordance with Part V of the Specifications of Highway

Bridges. Factor 1/3 represents an amount of

increase of subsoil deformations due to liquefaction, it is shown in Table 3 . Even when the sublayer which CAB is to be buried is stable against liquefaction, the factor 1/^ shall be considered when sublayers underlying

that sublayer are likely to cause liquefaction.

For seismic design of straight chain structures embedded in shallow ground like CAB, the ground

strains developed along CAB axis has the controlling effect. The ground strains during

earthquakes are contributed by several variables such as non-uniformity of subsurface

ground and effect of wave propagation; the standard ground strain was assumed as 1/100. Although the ground strains obtained by field observations are generally much smaller than

1/100, the value of 1/100 was adopted for Eg taking into account that significant cracks and subsidence associated with large deformations

of subsoils were frequently observed in past

large earthquakes

.

2 .

2

Seismic Design

Seismic force P developed in CAB's axial direction and the magnitude of joint length A (avoid uncoupling of joint) are

155
P - D X X rg (2) A -

«G ^ (3) where D: width of CAB in touch with surrounding soils per unit length

(m)

: length of CAB element (m)

rg ; frictional force between CAB and

soils per unit contact area (tf/m^)

«q; seismic ground strain represented by

Eq. (1) The seismic force P defined by Eq. (2) represents the force transmitted from soils to CAB associated with deformations of surrounding

subsoils. The joint length A by Eq. (3) represents the relative displacement between an

interval CAB element associated with the seismic ground strains. Note, such relative displacement has to be absorbed by joints without causing uncoupling of joint.

3) installation of pull-out prevention device

at the joint at sites where large ground

deformations are likely to occur.

3. DYNAMIC LOADING TEST OF CAB -ELEMENT

3 . 1 Objectives

Underground structures similar to CAB such as drainage U-ducts, water supply and sewage pipes

have sustained damages in past earthquakes.

The damage features of these underground

structures are characterized as:

1) Damase of Pipe Itself (refer to photograph 11

Longitudinal and transverse cracks were

developed, at times the pipe was crushed from

large deformation and movement of subsurface

soils during earthquakes.

2.3 Judgment of Liquefaction of Sandy Soil Lavers and Countermeasures for

Vulnerability for liquefaction at certain depth

may be represented in terms of liquefaction

resistance factor Fl- However, Fl gives

liquefaction vulnerability only at a specific

depth considered and is not necessarily a suitable index representing stability of the site associated with liquefaction. Therefore,

the liquefaction potential index Pl, which is designated as the integral of Fl from ground

surface to 20 m below the surface, was adopted

to evaluate the stability of the site.

It is stipulated in the manuals that the safety

against uplift of CAB be checked when Pl becomes larger than 15 and when CAB is installed in the sandy layer with Fl less than

1.0. The excessive pore water pressure vs. Fl

relation, which is used for evaluation of safety factor against uplift, is specified in

the manuals

.

Countermeasures for liquefaction should be conducted when the surrounding subsoils have high potentials for liquefaction (Pl > 15), and

possible movement of surface layer associated

with liquefaction is expected.

2 . 4 Structural Details in Seismic Design

For improving seismic performance of CAB, the following structural details are stipulated:

1) installation of joint at the points where

sudden change of ground condition occurs,

2) adoption of joint to absorb required joint

displacement given by Eq. (3),

2) Failure. Shift and Uncoupling of Pine Joint

(refer to photographs 4, 5, and 6) The pipe joint is generally more vulnerable

than the pipe itself and failure, shift, and uncoupling at pipe joints were frequently

observed. According to lessons learned through past

earthquake damages, it is considered important

to prevent future seismic damage of CAB by

improving the strength of its main body

including joints by providing pull-out pre- vention devices (stopper) at joints. Consequently, the purpose of these experiments verified the strength of CAB element and the effectiveness of the stopper.

3 . 2 Test Cases and Loading Condition

3.2.1 Test Cases

A series of four loading tests was conducted;

the purpose of and specimen used for each test

are described.

1) Test 1 In test, 1 the strength of CAB in its axial direction was studied. One side of the CAB specimen was pushed and/or pulled by a dynamic

actuator, the other side was fixed to a reaction wall as shown in Fig. 3. Fig. 4 shows dimensions and reinforcement of the specimen

used for test 1.

2) Test 2 In test 2, the strength of CAB's bending moment

around a vertical axis was studied. A dynamic

lateral force was applied by the actuator to the CAB specimen. The specimen was supported

by two hinges placed at both ends using a two leg jig as shown in Fig. 5. It produced

156
constant bending moment in the specimen. Fig.

6 represents the dimensions and reinforcement

of the specimen for test 2.

3) Test 3 In test 3, the effectiveness of the joint

stopper was studied. Three elements of CAB

were placed as shown in Fig. 7. Four stoppers

were set for connecting two adjacent elements.

One side of the end element was pushed and

pulled by the dynamic actuator; the other end

was fixed to the reaction frame. Fig. 8 represents the dimensions of the specimen for

test 3. The reinforcement of this specimen is the same as denoted in Fig. 4. Test 4 In test 4, the strength of joint in transverse

direction was studied. Three elements of CAB

were connected as shown in Fig. 9; the end of

two side elements are supported by hinges.

Dynamic lateral force was applied to the center

element so shearing force could be applied at

the joints. Fig. 10 represents dimensions of

the specimen for test 4. The reinforcement of

this specimen is essentially the same with that represented in Fig. 4. The specimens used in the loading test are full-scale models of actual CAB's constructed

as a pilot execution at one of the city roads

in Tokyo metropolitan area by the Ministry of Construction in 1983 and 1984. In the loading

tests, strains developed in the reinforcements

and at the concrete and relative displacement

between two adjacent elements at the joint were measured including the lateral dynamic force applied to the specimen. The strains were measured at 7 points as shown in Fig. 11. 3.2.2 Loading Conditions

The seismic force required for the seismic

design was obtained in accordance with the manuals. The lateral dynamic force applied to the specimens (refer to Table 4) had an

amplitude approximately three times as large as

the required seismic force. The loading of sinusoidal wave with frequently of 0.5 Hz and with number of cycles of 10 was reproduced by a dynamic actuator installed at the dynamic

structural laboratory, P.W.R.I.

3 . 3 Test Results

1)

Test 1 The strain of reinforcement vs. loading force relations were obtained as shown in Fig. 12. Fig. 12 shows that differences of measured

strains depending on the measuring points are very small. This implied that the specimen

developed uniform extension and compression.

Fig. 12 illustrates that the strains predicted,

assuming that full section of concrete resists

for the force, agree well with the measured

strains

.

The strains for reinforcement and concrete corresponding to the seismic force of ll.lt, expected during earthquakes, are only 20 /i, i.e., stresses developed in the reinforcements

and at the concrete are only 40 kg/cm^ , and 5 kg/cm^ , respectively.

The seismic performance of the specimen are shown in Fig. 13, i.e.,

1) no damage was developed up to the lateral

force of 15t.

ii) at lateral force of 15t, hair cracks were generated at the knockout, and they grew slightly in accordance with increase of

the lateral force.

iii) at lateral force of 35t, only hair cracks developed around the knock-out. The CAB specimen resisted without serious damage

for a force 3 times as large as the seismic force, which is expected to develop during earthquake.

2) Test 2 Fig. 14 shows the strain of reinforcement vs. loading force relation for test 2. As the case

for test 1, predicted strains assuming that

full section of concrete resists for bending

moment agreed well with the measured strains.

The strains developed in the reinforcement and concrete at force level of 21t (three times the force supposedly developed during an earthquake) are only 70 n, i.e., stresses of reinforcement and concrete are 150 kg/cm^ and

20 kg/cm^ , respectively. Only hair cracks appeared at the knockout as shown in Fig. 15.

3) Test 3 Fig. 16 represents the hysteresis curve of relative displacement between two adjacent

elements vs. force developed in the stopper at the joint. Because collision occurs at the instance when the relative displacement becomes

zero, increase of force level is observed in the hysteresis curve. Although some spalling of concrete occurred at

the joint due to collision, it was confirmed

that the stoppers work effectively to prevent pulling-out of one CAB element from the adjacent CAB element. Note, because the axial

force developed in the specimen was small, no cracks appeared in the specimen.

157
4) Test 4 The redundancy between two adjacent specimen

connected by tenon and mortise in the trans- verse direction is small (refer to Fig. 8(b)), few relative displacements occurred during

excitation. The strains developed in the reinforcement and the stoppers were small; no

cracks or damage occurred.

4. CONCLUSIONS OBTAINED FROM DYNAMIC LOADING

TESTS

From the results of the dynamic loading tests

of CAB- specimens , the following conclusions

were derived:

1) the seismic performance of the CAB specimen

for the four tests are summarized in Table

5.

For seismic force, expected during an earthquake, no damages were observed except

for minor hair cracks at the knock-out.

For the seismic force three times as large

as the expected developed seismic force, only hair cracks appeared at the specimen

in the vicinity of the knout-out. From

these results, it was confirmed that the CAB has sufficient strength to survive

seismic forces,

2) the effect of the stopper, for prevention

of pulling out at the joint, was

significant, and

3) the strain developed in the reinforcement

and concrete can be sufficiently predicted

by assuming that the full -section concrete

resists tension and bending moment.

5. ACKNOWLEDGMENT

This report presented the seismic design

procedures of CAB adopted in the Manuals for Design of CAB System (Draft) . It was issued by

the Ministry of Construction in 1985; the dynamic loading test was conducted to verify

the seismic safety of CAB. Cooperation to perform the dynamic loading

tests was received from the Road Bureau and the Kantoh Regional Construction Bureau of the Ministry of Construction, the Research Center

for Land Development Technology, and the Highway Industry Development Institute. The authors appreciate all concerned for their cooperation during this study. Special

appreciation also goes to members of the CAB Committee chaired by Prof. Masakatsu Koshi, University of Tokyo, for their valuable

discussions in formulating the seismic design

procedures and for conducting the dynamic

loading tests.

6. REFERENCES

1) Highway Industry Development Institute, Technical Manual of CAB Systems, January

1986 (In Japanese).

2) Kawashima, K. , N. Obinata, K. Gotoh, and

T. Kanoh, Investigation on Seismic Perfor- mance of CAB, Part 1, Preliminary Study

for Seismic Performance of Sandard Type

CAB, Technical Memorandum No. 2138, Public Works Research Institute, 1984 (In Japanese)

.

3) Kawashima, K. , et al.. Investigation on Seismic Performance of CAB, Part 2, Dynamic Loading Tests of CAB, Technical Memorandum No. 2259, Public Works Research

Institute, 1985 (In Japanese).

4) Kawashima, K. , N. Obinata, K. Gotoh, and

T. Kanoh, Damage of Sewage Facilities Caused by the Nihon-kai-chubu Earthquake

of 1983, Part 1, Damage Features of Sewage Facilities, Technical Memorandum No. 2244, and Part 2, Photographs Showing Damage, Technical Memorandum No. 2237, Public Works Research Institute, 1985 (In Japanese)

.

5) Kawashima, K. , N. Obinata, K. Gotoh, and

T. Kanoh, Seismic Damage of Sewage Pipes Caused by the 1983 Nihon-kah-chubu

Earthquake, Pressure Vessels and Piping

Technology, ASME, New Orleans, Louisiana,

U. S.A., 1985.

6) Japan Road Association, Specifications of Highway Bridges, Part V, Earthquake Resistant Design, 1980.

7) Iwasaki, T. , Soil Liquefaction Studies in Japan, Soil Dynamic and Earthquake

Engineering, Vol. 5, No. 1, 1986.

8) Katayama, T. and N. Sato, Ground Strain Measurements by a Very Densely Located

Seismometer Array, Proc. of 6th Japan

Earthquake Engineering Symposium, 1982.

9) Iwasaki, T. , Earthquake Resistant Design

of Underground Pipelines in Japan, Proc. of US -Japan Workshop on Seismic Behavior

of Buried Pipelines and Telecommunications

System, Tsukuba Science City, Japan, 1984.

10) Kawashima, K. , K. Tamura, and K. Aizawa, Finite Ground Strains Induced During

Earthquake for Application to Seismic Design of Underground Pipelines, Proc. of

U.S. -Japan Workshop on Seismic Behavior of Buried Pipelines and Telecommunications

System, Tsukuba Science City, Japan, 1984.

158
11) Iwamoto, T., N. Wakai, and T. Yamaj i

,

Observation of Dynamic Behavior of Buried

Pipelines During Earthquakes, Proc . 8th World Conference on Earthquake

Engineering, San Francisco, U.S.A., 1984.

12) Kubo, K. , M. Hamada, and R. Isoyama, Measurement of Permanent Movement of Subsurface Ground Due to Nihon-kai-chubu

Earthquake, Proc. of Earthquake

Engineering, JSCE, 1985 (In Japanese)

.

159
Table 1 Seismic Zone Factor v,

Zone A B C

Value of 1 .0 0.85 0.7

Table 2 Ground Condition Factor Vj

Group 1 2 3 4

Value of y, 0.9 1 .0 1.1 1 .2

Table 3 Liquefaction Effect Factor v.

Group Coefficient w. Condition Remarks

1 1 .0 P < 15

L

mu'/ L L

2 1 .5

P >15 and

CAB is installed within soil layers with ^ Fl<1

3 2.0

P >15 and

L =

CAB is installed within soil layers with ^^<1

]

Fl<1

J Fl< 1 u! C(

Cf

til::; C;

160
Table 4 Load and/or Displacement Subjected to CAB-Specimens

Case Estimated Seismic Effect (Load and/or

Displacement) Subjected to CAB during

Earthquakes

Load and/or Displacement Applied to CAB Specimens by This Experiment if) / 'A) Procedure for Loading Contro*

Static Loading Test Dynamic Loading Test 9

Test-1 11.1 t 20 t 35 t 3.2 Load Control

Test-T 6.8 t 10 t 21 t 3.1 Load Control

Test-3 Relative Displacement at Joint

between Two Adjacent CAB elements

± 10 mm

Compression 35 t Tension 4 tRelative Displacement

at Joint

1. ’0 mn —

Load Control (Static) Displacement Control

‘ Dynamic)

Test-.. 0.09 t 2 t 2 t 22.2 Load Control

Table 5 SunOTary of Loading Test Results

Case CAB-Specimens Subjected to The Load and/or Displacement Expected to Develop during Earthquakes

CAB-Specimens Subjected to 3 Times as Large as The Load and/or Displacement

Expected to Develop during Earthquakes

Test-1 No Damage Hair Cracks at Knock-out

Test-2 Hair Cracks at Knock-out Hair Cracks at Knock-out

Test-3 No Damage, but Slight Spalling-off of Concrete at Joint due to Collision

Test-4 No Damage No Damage

161
(b) Transmission Line Stored in CAB

Fig.1 Effect of CAB-System for Distributing Electricity

and Telephone ri

ci:

162 1
Fig. 2 Standard Cross Section of CAB

Specimen

A Fixing Wall for Jig

1 1

^7777 /////

(b) Side View

Fig. 3 Loading Procedure for Test 1

163

Specimen

2 Leg Loading Jig i

T Actuator

(a) Plan View

Load Transmission Jig

Objective Part of Specimen

Hinge o

(b) Side View

Fig. 5 Loading Procedure for Test 2

4 500

165

Fig. 9 Loading Procedure for Test 4

a_ an 505 245 245 505 _ I 500

500 1 _jL

75a L5o__ 1500

Fig. 10 CAB-Specimen for Test-4

167
Fig. 11 Measuring Points of Strains Developed in Reinforcing Bar

(SI - S7)

Fig. 12 Load vs. Strain of Reinforcing Bar

for Dynamic Loading Test with Frequency

of 0.5Hz (Test 1, refer to Fig. 11 for measuring points of strain)

168
E

rl5 m

(a) Micro Cracks Observed After Dynamic Loading with Amplitude of + 15t (f=0.5Hz)

A 1

1

1

1

1 — .! - -

1 B H

(b) Micro Cracks Observed After Dynamic Loading with

Amplitude of + 30t (f=0.5Hz)

Fig. 13 Dynamic Loading Test Results for Test-1

Fig . 14 Load vs. Strain of Reinforcing Bar for Dynamic Loading

Test with Frequency of 0.5Hz (Test 2, refer to Fig. 11 for measuring points of strain)

1 1

A r-H

1 ^-4-^ ~~r~

0 1 0

0 1 0 t IS A

Fig. 15 Micro Cracks Observed After Dynamic Loading with

Amplitude + 15t for Test-2

169
FORCE

DEVELOPED

IN

STOPPER

(t)

FORCE

DEVELOPED

IN

STOPPER

(t)

RELATIVE DISPLACEMENT D2 (mm)

(a) Dynamic Loading with Frequency of 0.3Hz

-5 1 -25 -20 -15 -10 -5 0 5 10 15 20 25 RELATIVE DISPLACEMENT D2 (mm)

(b) Dynamic Loading with Frequency of 0.5Hz

Fig. 16 Hysteresis of Force Developed in Stopper vs. Relative Displacement at Joint (No. 2 Joint)

for Test-3

170
^

' 0

Fig. 17 Spalling-off of Concrete Developed' at Joint

due to Collision (Test-3)

Photo 1 Installation of CAB (U-Shaped Duct)

171
Photo 4 Pulled-Out of Sewage pipe

at Joint (Nihonkai-chubu Earthquake)

Photo 5 Offset of Flume at Joint (Nihonkai-chubu Earthquake)

Photo 6 Offset of Drainage Flume

at Joint (Nihonkai-chubu Earthquake)

172
LARGE-SCALE BRIDGE COLUMN TESTS A PROGRESS REPORT

BY Edgar V. Leyendecker

William C. Stone Geraldine S. Cheok^ ABSTRACT

An experimental program to assess the performance of ful l-scale bridge

columns subjected to cyclic Inelastic loading has been undertaken at the National Bureau of Standards. The objectives ofthls program are to determine the adequacy of existing

California Department of Transportation

(CALTRANS) reinforcement provisions for highway bridges constructed In zones of high seismicity, and to determine the feasibility of reproductIng protype

behavior with 1/6-scale model

repi icates. A description of the on- going test program, as well as results from the first series of three 1/6-

scale model tests. Is presented.

1 . I NTRQDUC T .I..Qil A large portion of modern bridge

structures constructed In zones of high seismic activity are supported by bents consisting of one or more columns.

Nearly all present design codes for bridge columns seismic details have had their basis In the extensive research

done on building columns at relatively

small scale. Inelastic response of these bridge structures under seismic loading normally Involves plastic

hinging of the columns. Behavior Is consequently different In concept from that required of building frames, where

a design approach Is adopted to Insure that beams hinge before columns.

In an effort to summarize the current state-of-know I edge and to establish

research needs for seismic aspects of highway bridge design, the National

Science Foundation sponsored a workshop

on Earthquake Resistance of Highway

Bridges. The Proceedings, published In January 1979, contained recommendations

for future research, as well as state- of-the-art and state-of-practice

papers. Research recommendations were

called for in a number of areas.

A key recommendation was to perform

experimental and analytical studies

to determine the reserve capacities

of bridge components. Much of the considerable research work on column behavior has been done on relatively

small specimens and has been

extrapolated for bridges from tests of columns typically used in buildings. Bridge columns are larger

and lower stressed than building

columns; extrapolation from the

present wealth of building column

data may no be appropriate.

Therefore, work is needed to determine whether the behavior of small sections can be extrapolated to larger cross sections.

2. RESEARCH PROGRAM

A test program was developed using

the variables shown In figure 1. Specimens Include a flexural specimen

and two shorter specimens with

variable axial load. Tests are

planned at full and reduced scale. Two types of cantilevered bridge

columns have been designed based on full-scale 60” (1.5 m) diameter

bridge columns. The basic specimen

properties are as shown in figure 1; Type I) 30 foot (9 m) column(high

moment-to-shear ratio) with

continuous longitudinal reinforcement

through the footing for evaluation of transverse reinforcement performance

In the plastic hinge region (maximum

ductility factor) under varying axial

load. Type II) 15 foot (4.6 m) column (low moment-to-shear ratio) to investigate

shear performance of transverse

re I nf orcement.

^Structures Division, National Bureau of Standards,

Gaithersburg, Maryland 20899

173
3. FULL- SCA L E TESTS are noted for The facilities at the National Bureau of Standards (NBS) provide the

experimental means to perform large

scale structural testing. The basic

loading scheme is simple and Includes a combination of vertical and lateral

load. This is shown schematically in figure 2. Vertical load is applied

using an existing testing machine and lateral load Is applied using a reaction wall to move the test specimen

base. Although simple In concept, the test facility, shown In figure 3, is complex In practice. The structural laboratory Includes a 12 million pound

universal testing machine. Specimen

sizes of up to 58 feet in height and 8 feet In width can be accommodated In the machine. Lateral loads can be Imposed on the specimen by means of the newly completed 45 foot (14 m) high

react Ion wall. The first of three planned full-scale

test specimens 5-foot (1.5 m) diameter

and 30-feet (9 m) high) has been

fabricated, see figure 4. Testing Is planned for early July, 1986.

4. MO D.LL.J.LSXS Model tests are being conducted at 1/6- scale using both microconcrete and conventional aggregate concrete. These are being tested using the NBS TriDlrectional Testing Facility shown In figure 5. A model of the column Is shown In figure 5

.

Various stages of the model program are

shown in figure 6 (model of the 30 foot

(9 m) specimen). Note the fractured

vertical reinforcement. Four 1/6-scale models tested during

May of 1986. A set of three specimens

model I ing the parameters In figure 1 using microconcrete have been

completed. These are being repeated

using small aggregate concrete, but not scaled as m icroconcrete. Preliminary results from the

microconcrete tests are shown in figure

7. Only the tip deflection data are shown In the figure. Much more data have been collected but are still in the process of being analyzed.

The following points

the three specimens:

30-foot (9 m) model, 1000 kips (450

tons) vertical load

o spiral did not fracture

o vertical steel fractured at 8 times the yield displacement,

first eye 1

e

15-foot (4.6 m) model, 2000 kips (900 tons ) vertical load

o spiral fractured at 10 times

yield displacement, first

eye I e o vertical steel fractured at 10 times yield displacement,

third cycle

5. SUMMARY The current status of the NBS

research on the large-scale bridge

column has been briefly presented.

Preliminary model data have been reviewed. The full-scale

experimental program Is scheduled to continue through the spring of 1987 with the test of the first largescale specimen scheduled for July,

1986. Reports on the model tests are

due In October. Final reports on the comparison of full-scale and model specimens are planned for September

1987.

174
NBS

LARGE

SCALE

BRIDGE

COLUMN

TESTS

(Series

A)

o

175

Figure

1.

Summary

of

Test

Specimens

Schematic of Column under Test Reaction Wall and Testing Machine

Figure 3. Testing Facility for the Large-scale Bridge Project

177
Figure Large-Scale Specimen

178
179

Figure

5.

Tri-Directional

Testing

Facility

with

1/6-scale

Model

]

I

)

181
OUTLINE OF MANUAL FOR REPAIR METHODS

FOR CIVIL ENGINEERING STRUCTURES DAMAGED BY EARTHQUAKES

by

Toshio Iwasaki, Shigetoshi Kobayashi,

Takashi Uda, Yohtaroh Asai, Yasuyuki Koga, Yasushi Sasaki, and Kazuhiko Kawashima

ABSTRACT The Ministry of Construction completed a 5- year research project entitled "Development

of Repair Methods for Structures Damaged by Earthquakes," in March, 1986. The purposes

of this project is to develop a guideline

for inspection and measurement, assessment

repair, and overall evaluation methods of repairing, for structures and natural slopes affected/damaged by earthquakes. This paper

describes an outline of "Manual for Repair

Methods for Civil Engineering Structures

Damaged by Earthquakes, “ which is an important part of the results of the 5-year project conducted at the Public Works Research Institute and the Geographical

Survey Institute, the Ministry of Constructi on,

1 . INTR ODUCTION

Located along the Pacific Seismic Region,

Japan is one of the seismically disastrous

countries in the world and has often suffered from large earthquakes. The recent advancement in structural and earthquake

engineering, however, has enabled us to safely design and construct huge structures

such as long-span bridges, underwater

tunnels, high-rise buildings, etc. The research entitled "Development of New Seismic Design Methods for Buildings and Structures," was performed by the Ministry

of Construction in 1972 through 1977 using

lessons gained from the Tokachi-oki Earthquake of 1968 and the San Fernando Earthquake of 1971. This research has led

to the advancement of Japanese seismic

design codes.

As a result, catastrophic disasters due to entire collapses of buildings and structures

have decreased and the numbers of human fatalities have become comparatively lower during recent large earthquakes. It is considered, however, that minor damages and

partial failures may still break out in

future earthquakes, as observed in the Miyagiken-oki Earthquake of 1978 and the Nihon-kai -chubu Earthquake of 1983. In view

of this fact, post-earthquake measures

including the development of repair methods

for partially damaged structures have become significant concerns.

The Ministry of Construction initiated a comprehensive research project, "Development

of Repair Methods for Structures Damaged by Earthquakes," in April, 1981, and completed

it in March, 1968. This project developed post-earthquake measures by providing

procedures of inspection and measurement, assessment of damage extent, repair methods,

and overall evaluation for repairing

seismically damaged structures. This paper

introduces an outline of "Manual for Repair

Methods for Civil Engineering Structures Damaged by Earthquakes , " an important part

of the results.

In executing this project (between 1981 and 1985), the Technology Research Center for National Land Development established a Committee for Development of Repair Methods

for Civil Engineering Structures Damaged by Earthquakes (Chairman; Prof. Shunzo Okamoto) to discuss and examine the Manual,

with the commission from the Public Works Research Institute, the Ministry of Construction. At the same time, four Sub- committees, namely Sub-Committee on Earth Works (Chairman: Prof. Kenji Ishihara), Sub-Committee on Bridge Structures

(Chairman: Prof. Motohiko Hakuno), Sub- Committee on Underground Structures

(Chairman: Prof. Tsuneo Katayama), and Sub- Committee on Evaluation Methods of Repair

(Chairman: Prof. Keizaburo Kubo), were organized under the above Committee to prepare a draft Manual. Comprehensive

experiments and investigations were conducted at the Public Works Research

Institute and the Geographical Survey

Institute, the Ministry of Construction.

Research for each of the four Sub-Committees

Public Works Research Institute, Ministry of Construction, Tsukuba, Japan

182
are listed in Table 1 in chronological

order.

2. FLOW OF REPAIR AND RESTORATION

2.1 Flow of Repair and Restoration

Repair and restoration of seismically

damaged civil engineering structures shall

be correctly and immediately conducted

because it significantly affects restoration

activities of the public and stabilization

of national life. Repair and restoration

shall be done in accordance with the disaster plans by correctly understanding

the whole damage, and by consulting and exchanging information with related organizations and authorities. Flow of repair and restoration after an earthquake

may be classified into three stages as shown

in Fig. 2-1. The fundamental aims of the three stages are the following;

1) First Stage of Repair and Restoration

Inspect damage with emphasis on the important and/or critical facilities as early as possible, and also decide the strategy for repair and restoration. When

it is judged that a large secondary disaster

is likely to develop, conduct appropriate

treatment.

2) Second Stage of Repair and Restoration

Inspect damage of all facilities, and judge

the necessity of temporal repair and restoration based on the consideration

for large possible secondary disaster,

emergency for restoration, types and importance of the facilities, and the time

required for initiation of permanent

restoration. When it is judged that the

temporal repair and restoration is required, conduct it by the most appropriate

method considering priority and restoration

level

.

3)

Third Stage of Repair and Restoration

Determine the level required for permanent

repair and restoration based on importance,

location of damage, damage degree, difficulties of repair and restoration, and future plan. Conduct permanent repair and restoration by considering the future development and restoration plans of the

site.

2 . 2 Items to be Considered for Conducting

Repair and Restoration

1) First Stage of Repair and Restoration Conducting urgent damage inspection and urgent treatment in the first stage of repair and restoration; the following items should be considered.

a) Characteristics of main shock, aftershock, tsunami and weather

b) Characteri sti cs of damage area

c) Potential for secondary disaster

d) Damage of other facilities

2) Second Stage of Repair and Restoration

Determine fundamental strategy for repai

r

and restoration in the second stage; the following items should be considered.

a) Characteristics and degree of damage

b) Influence of secondary effect including

secondary disaster

c) Possibility of occurrence of secondary

effect including secondary disaster

d) Restrictions for conducting repair and restoration

Table 2-1 represents major items which

should be taken into account for the second

stage.

3) Third Stage of Repair and Restoration

Conduct permanent repai r and restoration;

the following items should be considered.

a) Future plan of the damaged area and

faci lity

b) Possibility for sustaining damage in future

c) Adjustment with related other

faci 1 ities The repair and restoration level shall be determined based on the possible increase of function of damaged facility, possibility

and appropriateness to restore the facility.

183
feasibility of repair and restoration, and required increase of seismic performance.

Fig. 2-2 shows a flow chart to select permanent repair and restoration.

3. REPAIR AND RESTORATION TECHNIQUES FOR RIVER, COASTAL AND SABO FACILITIES

This manual describes the restoration and repair techniques for seismically damaged

river, coastal and sabo facilities. River facilities include river dykes, water

embankments, sluice gates, etc. Coastal facilities are coastal dykes, water

embankments, detached breakwater, wave dissipation breakwaters, etc. Sabo

facilities are concerned with erosion

control works, landslide prevention works,

steep slope collapse prevention works, and slope failures including enormous slope

failures related to these facilities.

3.1 Fundamental Plan

1) Fundamental Plan for Restoration and Repairing of River, Coastal and Sabo

Faci 1 ities The fundamentals in the restoration of seismically damaged facilities is in temporarily protecting secondary disaster to

be caused by rainfall, flood, high water, aftershock, etc. and executing permanent

repair and restoration.

2) Flow of Repair and Restoration

Fig. 3.1 shows the flow chart from occurrence of an earthquake to investigation

and end of the restoration. This flow can

be roughly divided into three stages considering time, object and content. The first stage in the restoration is from the occurrence of an earthquake that caused

damage to understanding the outline of damages, and executing some immediate

treatment to prevent a secondary disaster.

The duration is roughly one day, through it depends on the area and the amount of damage. Next, in the second stage the damage is investigated in detail and temporary repair and restoration is executed

where required. Finally in the third stage permanent repair and restoration is executed

to regain the original function of damaged

faci 1 ities.

3) The First Stage of Repair and Restoration

In the first stage, the sites which need urgent treatment against possible secondary

disaster, are selected and the strategy for the inspection and restoration thereafter is considered. The urgent treatment in this stage is generally limited in a slight

action without essential construction.

4) The Second Stage of Repair and Restoration

In the second stage, the damage situation in the whole administration area is investigated in detail mainly by site reconnaissance taking a few days to a week. Based on investigation results, it is judged whether restoration should be conducted, and whether temporary repair and restoration is needed, that is, whether restoration is conducted under two stages of temporary and permanent repair and restoration, or under

only one stage of permanent repair and restorati on. Moreover technical methods for temporary and permanent restoration works are selected and the approximate amount of temporary and permanent restoration works is estimated. Based on such judgement temporary

restoration works are executed. For the judgement of the need for restoration works

and temporary restoration works, the damage

pattern and rank, the period until the permanent restoration is executed, and the expected magnitude of external load during

the restoration period must be taken into account. When temporary repair and restoration works should be executed, damage

scale, allowable construction period,

appropriate restoration level, technical methods, extent of the works and priority

order of construction sites are to determi ned.

5) The Third Stage of Repair and Restorati on

In the third stage, to select and to design

technical methods for permanent restoration

works, and to clarify damage mechanism,

several types of investigation, mainly

ground investigation, are performed. Based

on investigation results, permanent repair

and restoration works are executed. For the judgement of the restoration level, the

184
selection of suitable technical methods and appropriate extent for restoration works,

the original function of damaged facilities,

the damage scale and mechanism, and the possibility of future damage must be taken

into account.

3.2 Restoration of River Facilities

1) Damage Inspection and Judgement of River

Faci 1 ities The length of dykes and levees are particularly large among the river

facilities, it is necessary to investigate

their damages by car along a dyke administration road to grasp the outline and location of heavy damage. When restoration

is needed a longitudinal and cross sectional

survey is performed in addition to site reconnaissance.

In the site reconnaissance, the damage to the facilities is classified into several patterns and ranks through measuring the

extent, width, depth and length of the

crack, settlement, tilting, gap, etc. The classification is used for the judgement of restoration need and selection of technical

restoration methods. The pattern and rank classification as in Table 3.1 (a) (b) is proposed concerning earthen river dykes.

2) Repair and Restoration Methods of River

Faci 1 ities Several technical restoration methods used

in the second and the third stage are classified based on restoration purposes considering damage pattern. More over the fundamental idea for each method and the

points to be considered in the designing and construction are arranged for the use of engineers. Repair and restoration methods

for river dykes are classified and shown in Tables 3.2 and 3.3. 3.3 Repair and Restoration of Coastal

Faci lities Coastal facilities may be damaged by tsunami, earthquake motion, combination of these and high water after the earthquake

which induces secondary damage. As regard

to damages due to tsunami , damage patters

are classified by examining various damages

of the previous tsunamis, and tsunami forces

related to them are also classified. In the

inspection of the damage due to tsunami, not only the damage situation of respective

facilities but also the survey of innundated

area of tsunami is important.

In the restoration, most appropriate

technical methods must be selected considering the degree of the damage, and the restoration to the original level.

3.4

Repair and Restoration of Sabo

Faci lities When a failure of a natural slope occurs, a large amount of soil is piled on a river bed and after forms a damaged lake. When an inspection by a visual observation from ground or observation, videotape, and oblique photograph indicates that such dangers as a slope failure from cracks at nearby slopes, a secondary debris flow by a rainflow from the accumulated collapsed soil

at a river bed and a break of dammed lake, warning and evacuation must be considered.

Several viewpoints for the judgement in the investigation are arranged separating the cases into a mountain slope, a river bed, a landslide, and a steep slope. The viewpoints of the measure level of temporary and permanent restoration works

are described. The technical methods concerning those facilities are arranged and classified considering the feature of the phenomena, and the applicability of these

methods is also summarized.

4. REPAIR AND RESTORATION TECHNIOUES FOR ROAD FACILITIES

4.1 Fundamental Strategy for Repair and RestorTtion of Road Facilities' ~

Post-earthquake functions required for roads may consist of connectivity, reliability,

mass transportation and safety as shown in Table 4.1. Although functions required for roads depend on the magnitude of earthquake

damage and characteristics of the area damaged; they change principally in the sequence as connectivity - reliability - mass transportation - safety in accordance

with time after the end of an earthquake.

Therefore, it should be considered when determining the target of repair and restoration that post-earthquake demand

required for roads change with time.

185
4.2 Urgent Damage Inspection and urgent Treatment

Main targets for urgent damage inspection

are the fol 1 owing

;

1) Is the road passable?

2) Investigate damage, especially critical damage,

3) Investigate for possible large secondary

disaster. Table 4-2 represents major check points.

The urgent damage inspection shall be conducted in accordance with Fig. 4-1. 4.3 Repair and Restoration of Highway

Bridge

1) Temporal repair and restoration

Table 4-3 summarizes major damage patterns

of bridges observed in past earthquakes. Among these damage patterns are pier

failures, failure of seat concrete, and failure of main members of superstructure.

They are considered the most susceptible

which are most likely to to result in bridge

failure. Consequently, damage inspection

for temporal repair and restoration shall be done for these three main items, and subsequently conduct inspection .in detail.

The damage degree for temporal repair and restoration shall be judged based on the

result of damage inspection.

i) Damage Judgement in View of Bearing

Capacity

A : No Damage - No special damages are detected

B : Slight Damage - Damaged, but not influential for short term service

C : Medium Damage - Influential damage, but structure may be used for short-term service

unless progress of damage due to aftershock

and live load is not developed

D : Critical Damage - Possible superstructure failure is expected due to damage

E : Falling-off - Falling-off superstructure is developed

ii) Damage Judgement in View of Passable Condition

a : No Damage - No special damages are detected

b ; Passable With Care - Damaged, but passable with care

c : Unpassable - Badly damaged, and unpassable

Fig. 4-2 is an example how damage judgement

in view of bearing capacity is made based on simple informations detected from out-view

of the facility, i.e., this example

represents a check point for judging the degree of damage to reinforced concrete pier which sustained bending failure at its base.

Fig. 4-3 represents a flow chart for required traffic restriction as well as temporal repair and restoration based on damage judgement in view of both bearing

capacity and passable condition. Table 4-4 summarizes major temporal repair and restoration methods.

2) Permanent Repair and Restoration

The damage degree for permanent repair and restoration shall be judged based on damage investigation as

A : No Damage - No damage or minor damage which does not influence function of bridge

from the view of long-term service

B : Slight Damage - Damage which does not affect bearing capacity

C : Medium Damage - Damage which affects bearing capacity

D : Critical Damage - Damage which significantly affects bearing capacity

For selecting the permanent repair and restoration method, time required, cost, workability, availability of construction materials and out-view aspect after repairing shall be considered. Table 4-5

186
represents permanent repair and restoration

method of reinforced concrete pier which

sustained bending failure at its base.

4.4 Road Embankments

This article discusses road embankments

which include the road surface, the embankment, the slope of the embankment and

the drainage facilities for surface water.

The paper describes the purpose, content and method of the inspection and judgement in the second and third stage of restoration.

The fundamental idea is the same as described in Chapter 3, River, coastal and Sabo facilities. Art. 3.1 Fundamental plan.

To judge appropriate restoration levels, technical methods, aerial surveys and priority order for temporary restoration, is necessary to classify the damage pattern and

rank which governs the serviceability and stability of the road embankment considering

the location, depth, and magnitude of cracks

and settlements of the road surface, embankment slope and slope protection works.

The classification of the damage pattern and

rank is proposed in Table 4.6 and Table 4.7. For temporary restoration it is necessary to consider the damage degree, the period

length until the full restoration is completed and the expected function of the

road during the restoration works.

The design level of permanent restoration

works must be determined considering the previous functions of the facilities, the importance of the road, the influence of the

damage, the past damage history, the future

damage possibility, etc. The purpose of technical methods for temporary restoration works can be classified into the protection of an embankment failure and the security of transportation. The purposes of these restoration works is classified into security of transportation function and the measure of stabilize embankments. The idea

of the applicability of the methods and the attentions in design and instruction are summarized. Table 4.8 and 4.9 show the classification of restoration methods of road embankments.

4 .5 Natural and Out Slope

This article addresses cut slopes, natural slopes, and slope protection structures concerning these slopes. The purpose, content and method of the investigation and judgement in the second and third stage in restoration is described as in road embankments in 4.4 In the investigation the collapsed soil and rock piled on the road surface from the slope and damage behavior

and stability degree of the slope must be given attention. To judge the appropriate

restoration level technical method, aerial extent, and priority order, it is necessary

to classify the damage pattern and rank, which governs the serviceability of the road and the stability of the slope, through the damage behavior such as swelling, crack movements, etc. and geomorphol ogi cal and geological condition of the slope. The need and procedure of the restoration must be judged considering the presence of a detour, importance of the road, damage degree and construction period. The classification of damage pattern and rank is proposed considering thickness, relative depth to a supporting ground, assess if falling rocks are predominant and the scale of the failure. The design level and appropriate

technical method in the permanent

restoration works must be determined considering the overall stability of the slope before damage, damage scale, damage mechanism, and the possibility of future

damage.

The purpose of temporary restoration works

can be classified into urgent security of transportation and protection of the secondary disaster. The purpose of permanent restoration works is classified by the types of restraint and protection

phenomena. The idea of the applicability of methods and attentions in design and construction are summarized for use.

5. REPAIR AND RESTORATION TECHNIQUES FOR SEWAGE PIPELINES

5.1 Scope of Description

Part 4 of the Manual describes the repair

and restoration techniques for sewage pipelines. Sewage facilities are composed

187
of pipelines, machinery, electrical, and building facilities. In these facilities, independently located facilities such as treatment station or pump station are comparatively easy to investigate or to repair their damage because most exist above ground surface. On the contrary, since pipeline facilities are buried beneath

ground surfaces having long total length spreading in wide area, they tend to be damaged by the direct effect of ground deformation and furthermore, once they are damaged by earthquakes, it is not easy to determine the damaged portions, rate of damage and restoration of their function.

In this part of the manual, information

describes the techniques for repairing the seismically damaged sewage works pipeline

faci 1 i ties. Since the installation of modern sewage facilities has had a short history, no large-scale sewage facilities in each large cities has suffered from earthquake damage.

However, since the Niigata earthquake in 1964, those sewage facilities in several cities with population between the tens to few hundred thousand had suffered damage due

to several earthquakes.

Among these earthquakes, the experience

obtained during repairing works after the Miyagiken-oki Earthquake (1978, M = 7.4) and the Ni honkai -chubu Earthquake (1983, M = 7.7) were used in developing the technical

manual for repairing works described in this

part of the manual

.

Part 4 of this Manual is composed of two parts, one is the lessons learned from damaged sewage pipelines from past earthquakes and the other describes the repair and restore techniques for sewage pipelines damaged by earthquakes. The table

of contents of Part 4 is included in the Appendix of this report, and the basic idea

of the contents follows.

5.2 Damage of Sewage Pipelines Due to Past Earthquakes

Outline of damage of sewage facilities due

to the past earthquakes since Niigata Earthquake is summarized in Table 5.1.

Intensity of seismic ground motion which

starts to cause damage to sewage facilities

is 4 in DMA scale. And there are lower bound in the earthquake magnitude varing with the epicentral distances which caused damage to sewage facilities as shown in Fig

5.1. Weak ground, adjacency to the rigid structure such as manhole and shallow

embedment tends to cause pipeline damage during earthquakes. In case of the occurrence of liquefaction of the ground where pipelines are buried, pipelines may suffer severe damage such as snaking deformation, uplift, uncoupling and/or shift

at pipe joints and in some cases the pipelines may be filled up by the sand flowed in from surrounding liquefied layer.

Fig. 5.2 shows an example of damage caused

by liquefaction. From these damage experiences during past earthquakes, it is learned that there are sites where post-earthquake inspection

should be carried out in detail and that there are several typical damage types of pipelines due to earthquakes. These lessons are part of the Manual.

5.3 Inspection and Investigation Methods

for Seismically Damaged Sewage Pipelinei~

After an earthquake, restoration work is implemented step by step in three stages which are urgent, temporal, and permanent repairing stages. At the first stage of repair and restoration, it is necessary to grasp whether the damage of sewage pipelines

may or may not causes secondary disaster

such as depression of road pavement or inundation from the result of urgent inspection to perform necessary treatment at the severely damaged portions. At this

stage, it is important to perform the inspection at whole sites of facilities as soon as possible. The inspection is mainly

implemented by observation from the ground surface without using equipment nor devices because of the shortage of time, equipments,

and devices. The key during this inspection

is to determine if there are deformations in ground surface at damage-prone sites as described in the previous section.
At the second and third stage of repair and restoration, it is possible to investigate

the pipelines from inside of manholes or pipelines as the consumable time for investigation becomes longer. At these

stages, it is necessary to clarify the

locations of damaged portion, original structure and buried condition with their

damage degree by the investigation so that

temporal or permanent repair works can be carried out according to the damage

condition.

The investigation method for this purpose is classified into two methods; one is the

direct investigation method and the other is the indirect investigation method. The

direct method is to investigate pipeline

damage by observation or measurement

directly using devices usually from inside

of pipelines. When the inner diameter of pipeline is too small for the investigator

to enter, it is effective to use a TV camera

mounted on a skid or self-driving crawler to obtain information about pipeline damage.

The indirect method is used for investigation of pipeline damage when it is difficult to enter the pipelines. The generally used indirect method is to determine damage of pipeline judging from discontinuity of fluid flow. For this purpose of investigation, measurements are employed such as the observation of the leakage of poured water from manhole or leakage of smoke injected from manhole or measurement of water flow at two points in the pipe. Applicability of these investigating methods

are summarized in Table 5.2. Considering the complex actions involved in earthquake investigations which are generally performed by many persons, it is effective to have given the unified

standards for judgement of damage and for reporting format of the investigation

results. Therefore in this manual,

guideline for evaluating pipeline damage and reporting the damage are proposed.

5.4 Repairing Method of Sewage Pipeline

Damage

When damage of pipeline facilities is found

by the investigation, damage is classified

into one of three categories of repair works —to implement urgent -treatment, to repair temporally and to repair permanently according to the degree of functional damage

and the possibility of causing secondary disaster to neighboring facilities.

The purpose of urgent treatment is to prevent secondary disaster by controlling

the traffic at the depressed point of road

or by pumping out the overflowed water.

The temporal repair is implemented to recover the pipeline function temporarily

until the completion of permanent repair. This category of repair includes removing

sand which flowed into the pipeline or manhole. High pressure cleaning mobile and vacuum cleaning mobile are used to remove

sand. The method for the permanent repair is selected from the investigation on the damage form and the damage degree. When performing permanent repairs, one should consider the existing life of the pipeline system and if the pipeline system

is expected to be expanded. Considering

these conditions, basic policy of repair works to recover the facility function

should be decided whether to repair

pipelines as they were, original-level

repairing, or to repair pipelines that

result is a significant improvement to this performance. Following the basic policy,

the repairing method should be selected by assessing the period of repair works and implementation circumstances besides the above mentioned damage form and damage

degree. Pipeline facilities are composed of various type of construction such as manhole, well,

small diameter service pipe, large diameter

pipe of trunk line, etc. The repairing methods for these structures are divided

into two groups, 1) manhole and 2) pipe. They are shown in Table 5.3. These repairing methods are classified into two groups by considering the purpose of works. One is to prevent leakage and the other is to recover the strength. The characteristics and applicability of each methods are introduced in the Manual.

189
Actual example of repair methods for various damage form and degree are shown in Table

5,4, which were employed to repair actual damage of pipelines caused by recent earthquakes.

6. SUMMARY

This paper introduces an outline of "Manual for Repair Methods for Civil Engineering

Structures Damaged by Earthquakes." It is a result of the Ministry of Construction's

project "Development of Repair Methods for Structures Damaged by Earthquakes" performed

1981 through 1986, Although this Manual is still a draft and is not a mandatory

regulation, it can be used as a reference

guideline when practical engineers conduct

repair works for seismically damaged

structures such as river engineering

structures, coastal structures, highway structures, and sewer pipelines, etc.

7, ACKNOWLEDGMENTS

The research project was implemented in "The Committee for Development of Repair Methods

for Civil Engineering Structures Damaged by Earthquakes," established at the Technology

Research Center for National Land Development, with a commission of the Public

Works Research Institute. The authors of this paper express their sincere appreciation to Professor Shunzo Okamoto

(Chairman of the above Committee), Professor

Kenji Ishihara (Chairman of Sub-Committee on Earth Works), Professor Motohiko Hakuno

(Chairman of Sub-Committee on Bridge Structures), Professor Tsuneo Katayama

(Chairman of Sub-Committee on Underground

Structures), Professor Keizaburo Kubo

(Chairman of Sub-Committee on Evaluation

Methods of Repair), committee members, and secretariats. The authors express sincere

thanks to individuals at the Public Works Research Institute and the Geographical

Institute who have provided valuable investigations through experimental works

and field surveys.
Table 1

Research

Items

of

Four

191
192
Table 2.1 Items To Be Considered in Executing Temporal Repair and Restoration

Items Examples Damage Degree o Damage of facilities not only administrated but nonadministrated

0 Functional damage of the area

0 Degree of degradation of structural performance of

facilities o Stability of damaged structure o Influence of functional damage

Probability of Occurrence

of Large Secondary Disaster o Factor depending on human activities 0 Factors depending on aftershock, tsunami, rain-fall etc.

Influence of Secondary

1 Disaster

1

0 Area affected by secondary disaster 0 Damage potential such as number of residence inside the affected area

Restriction for Initiation of Temporal Repair and

Restoration

o Restriction from social aspect

0 Restriction from approach to the site 0 Restriction from materials, instrument and operaters

0 Adjustment with other restoration

Restriction for Completion

of Temporal Repair and

Restoration

i

o Restriction from natural condition such that temporal repair and restoration be completed before

rainy reason

o Restriction from social condition

193
Third

Stage

of

Fig. 2-1 Flow of Repair and Restoration of Seismically Damaged Facilities

194
195

Fig.

2-2

Classification

of

Permanent

Repair

and

Restoration
Table 3.1 Classification of Damage Pattern and Damage Rank of River Dyke

(a) Classification of Damage Pattern

Damage

Pattern Schematic Figure Form of Damage

Type I Slope failure, collapse, cracks, and faulting are limited at the dyke shoulder.

Type II Sliding failure or longitudinal cracks and fault- ing reach the central part of dyke crown.

Type ir

Transversal cracks and faulting occur in the dyke.

Type III Failure reaches to the subsoils, and the dyke

does not show the origmal shape.

Type IV Uniform settlement deforms the dyke with

keeping the shape to a certain extent.

TypeV —T1

Settlement and cracks occur in dykes behind

structures.

(b) Classification of Damage Rank

Damage Pattern Damage Rank Description of Damage Rank

I & w ^ 15 cm, or 5 > 20 cm

y w < 15 cm, and 5 < 20 cm

II

CL w > 30 cm, and 5 > 50 cm

0 w < 30 cm, or 6 < 50 cm, and not equals 7 7 w < 15 cm, and 5 < 20 cm

ir

a w > 30 cm and 5 > 50 cm

0 w g 30 cm or 5 < 50 cm, and not equals 7 7 w < 1 5 cm and 6 < 20 cm

III a -

IV

a S > 100 cm

0 50 cm g S < 100 cm

7 S < 50 cm

V

a S ^ 50 cm

0 20 cm < S < 50 cm

7 S < 20 cm w : Crack width, 5 : faulting, S: settlement

196
Table 3.2 Temporary Restoration Technical Methods

Main Purpose Methods

Prevention of Rainfall Seepage San filling. Excavation and filling

Securing of Embankment

Might Sand filling. Excavation and filling. Sand-bag piling. Temporary coffering

Prevention of Leak through DykeSand filling. Sand-bag piling. Excavation and filling. Protection of front slope with good material. Sheet

piling at front slope HWL

Prevention of Leak through GroundSheet piling at front slope toe for protection of ground

water seepage

Erosion Protection Repair

of Damaged Rivetment

Sand-bag piling. Concrete filling

Securing of Dyke Stability Sand-bag’ piling. Sheet piling for earth support. Counterweight fill

Repair of Crack and Gap of Concrete Dyke

Sand-bag piling. Concrete filling. Temporary coffering

Table 3.3 Permanent Restoration Technical Methods

Main Purpose Methods

Securing of Dyke

Resistance Excavation and filling

Prevention of Leak

through Dyke do

Prevention of Leak

through Ground Sheet piling for protection of ground water seepage

Erosion Protection Rivetment

Securing of Dyke Stability Counter weight fill. Ground improvement

Securing of Concrete Dyke Stability Reconstruction, retrofitting and repairing of retaining wall

197
Fig. 3.1 Working Flow from Earthquake Occurence to Inspection and Restoration

198
Table 4.1 Post-earthquake Function of Road

Situation Definition

Connectivity Situation to be connected from a specific point to a

destination even if detour due to damage between

the two points is needed.

Reliability Situation able to go certainly to a destination within

a desired time with safety tolerable after earthquake

Mass Trasportation Situation able to serve for the same mass transportation

as the road had although reliability is not gueuranteed

Safety Situation able to service for transportation with the same safety as the road had

199
Table 4.2 Major Check Points for Urgent Damage Inspection

Facilities Check Point

Flat Road Large Subsidence of Road Surface

Road Low ~ High Embankment

Large Subsidence of Road Surface, Large Subsidence and Failure of Embankment

Natural and Cut

Slope Huge Failure of Slope, Large Fallen Stone on Road, Large Failure of Road

General Fall-off of Bridge

Slipping off and Gap of Hand Rail

Road Itself Bridge From the Surface Sudden Change of Leveling

Large Opening, Upheaval and Gap at Expansion Joint

Side Superstructure Sudden Change of Deformation and Leveling View

Substructure Large Subsidence, Tilting, Large Cracks and

Spalling-off of Concrete

Tunnel Large Slope Failure near Tunnel, Large

Spalling-off of Covering Concrete

w Common Duct Uplift to Ground Surface, Critical Damage

of Storing Facilities

Culvert and Underground

Pedestrian Pass Large Subsidence on Road Surface

Pedestrian Bridge Falling-off, Critical Failure of Pier

Way-side Facilities Collapse of Building on Road

Secondary Disaster caused by Damage of Road Facilities

Other Exclusive Use Facilities Influence of Damage of Exclusive Use

Facilities on Road

Others Critical Innaduation, Tsunami, Fire, etc.

200
Table 4.3 Damage Pattern of Bridges

Foundation I

Pile Foundation

Direct and Caisson Foundation

1

23

Pier Reinforced Concrete ~

Pier 4 — 5

Substructure

Steel

Pier Abutment 9

SuperI Concrete 10

structure ' Steel 11

Bearing Support

I— 12 — 13 — 14 — 15 — 16 — 17 —18 —19 — 20

Failure of Pile Top, Residual Displacement

Residual Displacement

(At Bottom of Pier) Crack and

Spalling-off of Concrete, Buckling and Cut of Reinforcing Bar, TUting

of Pier (At Cut-off Point) Crack and Spalling- off of Concrete, Buckling and Cut of Reinforcing Bar, Tilting of Pier (At Intermediate Height) Shear Crack and Spalling-off of Concrete, Exposure of Reinforcing Bar, Tilting of

Pier (At Sudden Change Position of Cross Section and Construction Joint) Crack and Spalling-off of Concrete

Buckling and Cut of Reinforcing Bar,

Tilting of Pier (Flame Structure) Crack and Spalling

off of Concrete, Buckling and Cut of Reinforcing Bar

Local Buckling, Cracks, Uneven Outplane Residual Deformation, Tilting of Pier Crack and Spalling-off of Concrete, Buckling and Cut of Reinforcing Bjut, TOting and Subsidence Crack and Failure around Support

Residual Deformation and Failure of Members around Support Deformation and Failures of Support and Sole Plate Deformation and Failure of Stopper

(Stopper of Upper Support, Side Block of Lower Support)

Pulling-out, Blending and Cut of Anchor

Bolt Looseness and Cut of Set Bolt (Pin and

Roller Support)

Pulling-out of Pin Cap, Cut of Cap (Pin and Roller Support)

EhiUing-out of Roller and Rocker

(Roller and Rocker Support) Deformation and Failure of Up Lift Stopper for Preventing Up-lift Crack and Failure of Mortar under Support

Crack and Failure of Seat Concrete

Expansion Joint

Other

— 21 Falling-off of Joint Filler — 22 Failure of Face Plate —23 Cut of Joint Rubber — 24 Failure of Anchor Concrete — 25 Opening and Contact of Joint — 26 Gap — 27 Failure of Hand-rail, Damper etc

201
Table 4.4 Major Temporal Repair and Restoration Methods for Seismically Damaged Bridges

Purpose Repair and Restoration Method

Prevention of Progress of Damage

a) Injection of Epoxy Regin

b) Correction of Deformation, Installation of Stiffner or Temporal Members

c) Installation of Falling-off Prevention Device

d) Temporal Support of Superstructure

Safety for Transportation and Pedestrian

a) Covering Work on Road Surface including Expansion

Joint

b) Correction of Gap

c) Temporal Repeiir of Hand Rail with Use of Steel Pipe

202
CL $

Design

by

Ductility

Analysis

Ultimate

Stage £0Z

Hybrid Hybrid Normal Hybrid Hybrid Hybrid

Yield

Stage

Normal Hybrid Hybrid

2/3

Es

for

Reinforcing

Bar 1/3

Es

for

Original

Reinforcing

Bar

1/3

Es

for

Original

Reinforcing

Bar

Ignore

Original

Section

ClC _c QE0 c 4- T3 3 £

:<S c ccnc 0 0 c £o2: >» >v X <2

'C0 c 0 = 5 c 2 .2 0 2 0 0 '-w 2 ^ 0 ’-w c3 is c ^ c 9 c ^ —.

CQ

T3 OJ 2 O 2 0) 5 CJ 0 Q 0) c c c S 0 0 0 c a C u a co '5b

o> -o

a; ’5b -o0) -o £

Qi 0>M Qci -2 o >i ol >. <2 c2 0a c

'£ 1 0a c£

“£0) c

'£ Cd 55 CjQ 05 05

•o >> >. >» > >» c o -Q 0 -Q n c c CiC Q£ _o C c _0 _C _C a 5 0) 5 0

'Z0> OJ 01 cS 0 0

•2, 0 0 0 a u U U 0 C0) CO u CO T3

a.

OJ C

oj "1

DC cc . 2P cu u. 55 £Q

OJ -

lac .= 0

<u <1> tUD C3 0> oc £ QQ

"o Q > 1 c 3 « ^ £

OJ '£

•a £ > c2 .2 5

'£

OJ c D 0 otf .2 <2 £

CQ 0 [ < ffl <

203

2)

Hybrid:

Original

damaged

section

can

be

treated

to

work

for

load

with

new

section.
Table 4-6 Classification of Damage Pattern and Damage Rank of Road Embankment on Flat Ground

(a) Classification of Damage Pattern Damage

Pattern Schematic Figure Form of Damage

Slope failure, collapse, cracks and Type I faulting are limited at the embank

ment shoulder.

Sliding failure, crack and faulting Type II reach the central part of embank- ment crown.

Failure reaches to the subsoils, and Type in the embankment does not show the

original shape. Uniform settlement deforms the Type IV /' embankment with keeping the shape to a certain extent.

Type V XU Settlement and cracks occur in in fills behind structures.

(b) Classification of Damage Rank Damage Pattern Damage Rank Description of Damage Rank

I

(3 w ^ 15 cm, or 5 ^20 cm

y w < 15 cm, and 6 < 20 cm

II a w > 30 cm, and 6 > 50 cm

/3 w ^ 30 cm, and 6 ^ 50 cm

III a

IV

P s ^ 50 cm

7 s < 50 cm

V

0

s ^ 20 cm

7 s < 20 cm w : crack width, 6 : faulting, s : settlement

204
Table 4.7 Classification of Shape and Damage Rank of Road Embankments

on Inclined Ground

(a) Embankment Shape on Inclined Ground Embankment of two

sides slope Embankment of one side slope Cutting and Embankment

Damage ranks are specified as the damage pattern and rank of the embankments on in- clined ground are almost coincidental.

(b) Classification of Damage Rank Damage Rank Description of Damage Rank

Oi Embankment slides completely, and road surface and shoulder disappears.

13 Embankment deformation causes cracks and faulting of the road surface. y No damage or hair crack on the road surface with no damage to the embankment.

205
Table 4.8 Temporary Restoration Technical Methods

Main Purpose Methods

Protection of Embankment failure caused by

Rainfall Seepage

Sand filling, Asphalt sealing. Sand-bag piling. Asphalt curb. Vinyl sheet. Temporary drainage

Protection of Embankment Failure caused by

Traffic Load

Filling, Sand-bag piling. Sheet piling for Earth Support, Wooden pile

Securing of Traffic (Securing of Width of Road, Securing of Detour)

Filling, Paving, Sheet piling for Earth Support, Pier or Temporary Bridge, Detour

Securing of Traffic Safety (Treatment of Gap Settlement)

Run-off, Sand filling. Asphalt Patching

Table 4.9 Permanent Restoration Technical Methods

Treatment of Rainfall Drainage ditch. Asphalt curb

Securing of Road

Function

Filling, Paving, Asphalt seeding. Injecting asphalt into cave

Securing of Embankment Stability when Embankment is Problematic

Filling, Retaing wall. Grating Crib Works

Securing of Embank- ment Stability when

Ground is Problematic

Counter weight fill. Ground Improvement, Underground

Drainage, Sheet piling

206
Fig. 4-1 Flow of Urgent Damage Inspection and Urgent Treatment

207
208

Fig.

4-2

Damage

Degree

of

Reinforced

Concrete

Pier

which

Sustained

Bending

Failure

at

Its

Base
-S'ocS

PeiTBAiicot Rep*ir

Fig. 4-3 Flow of Damage Judgement for Temporal Repair and

Restoration of Bridge

209
Table 5.1 Summary of Damage of Sewage Facilities due to Major Past Earthquakes

Name of Earthquake, Magnitude

(M), Date Area

Seismic

Intensity (JMA

Scale) Agency

Damage to Sewerage System

Sewer Pipe System Pumps and Treat- ment Plant Kanto Earth- quake; Tokyo VI Damage to sewer pipe system was relatively small compared with

those to other under- ground structures. Of toted extension of 180 km, about 250

places mainly within boundary of downtown

area and uptown area.

Minor damage to treatment plant was

observed, but not

so serious as to affect normal

operation.

Niigata Earth- quake; M = 7.9 June 16, 1964

Niigata City V Serious damage due to liquefaction was

observed. About 70%

of sewer pipe system of 35 km in total extension was damaged due

to rising, collision be- tween manholes and

pipes, disalignment, etc. About 90% was damaged in monetary value.

Of 15 facilities, 11 facilities were damaged due to de- struction of setti ing basins and rise of pumps caused by

liquefaction. Functions of 8 facilities were lost due to

serious structural damage.

Tokachi-oki Earthquake; M = 7.9 May 16, 1968

Hakodate

City V Damage due to lique- faction was observed in reclaimed area. Rise was observed.

No damage.

Muroran

City IV Of 5.5 km in total extension, portion

for 750 m laid in weak ground was set- tled.

Minor damages to two pumps were

observed.

210
Table 5.1 Continued Name of Earthquake, Magnitude

(M), Date

Area

Seismic

Intensity (JMA

Scale)

Damage to Sewerage System

Sewer Pipe System Pumps and Treat- ment Plaint

Miyagi-ken-oki Earthquake; M = 7.4 June 12, 1978

Sendai

City V Damages were observedin the area of weak

ground and newly

created housing site, but functions of sewage system in these

areas were maintained. Of total ex- tension of about 690 km, branch pipe system for about 630 m in service was

damaged. Serious damage was sufferred by sewer pipe system

not in service at the time of earthquake.

Of 11 pumping

equipments, functions of 9 pumps

were lost due to power failure and damage to pressure

pipe.

Siogama

City V Of total extension of about 27 km, portion

for about 700 m laid under weak ground

(reclaimed land) was

damaged.

Damaige was minor, and no failure of function was

reported.

Nihon-kai- chubu Earth- quake; M = 7.7 May 26, 1983

Akita

City V Damage to sewer pipe system due to liquefac- tion was reported. Of total extension of 286 km, relocated portion for 1.7 km

and 93 manholes were damaged.

Damage due to

to liquefaction was

reported. For

example, rise of puming equipment

at Tsuchizaki and

rainwater sedimentation basin were

observed.

Noshiro

City V Serious damage dueto liquefaction was

reported. Damage was

observed as to re- located portion for about 8 km of total extension of 60 km.

No damage was ob- served as to dis- posal plant under

construction.

211
Table 5.1 Continued Name of Earthquake, Magnitude

(M), Date Area

Seismic

Intensity (JMA

Scale)

Damage to Water Supply System

Buried Pipe System Pumps and Treat- ment Plant

San Fernando

Earthquake; M = 6.6 Feb. 9, 1971

San Fernando

Los Angeles

V-VI Dam2ige to pipe system

considered to be caused by the change

in the condition of ground resulting from

dislocation of geologic formation was re- ported. About 3.8 km

of pipe system in total was damaged (Restored later).

No serious damage

was reported.

212
Table 5.2 Summary of Investigation Method

Investigation Method Feature and Applicability

I. Direct Method

I-l Observation from Inside Applicable to large-diameter pipe, man-hole and small-diameter pipe which is adjacent to man-hole

1-2 Direct Measurement

from Inside Method to measure the joint-opening, crack width;

Applicable to large-diameter pipe and man-hole only; Necessary to shut off the flow and take measure to avoide

the accident by noxious gas or lack of oxigen.

1-3 Survey Survey of ground surface where pipe is hurried or the level of man-hole tops; survey of pipe axis from inside by using

special! devices such as laser; Applicable to large and small diaimeter pipes, and possible to investigate precise dislocation.

1-4 Trailed TV Camera Monitor the damage of pipe by using TV camera which is trailed inside the pipe; Measurable of the location of damage

and degree of damage by video-sealer; Applicable to pipes which is larger tham 20cm in diameter; There are two types of camera, one of which has only fore- sight and the other has fore-sight and laterail-sight. 1-5 Self-crawling TV

Camera Same as the “1-4 Trailed TV camera” but not necessary to trade the camera. Possible to investigate pipes inside from a man-hole.

1-6 Inserting TV Camera

(for service pipe) Monitor the damage of service pipe by TV caimera which is mounted at the end of snakable rod and inserted from well; Applicable to larger pipe than 10cm in diameter.

1-7 Special Water-proof Camera

Moniter the damage of pipe by photo taken by the special

i water-proof camera; Applicable to large-diameter pipe and

even to small-diameter pipe if the camera is mounted on

skid or crawler of TV camera.

1-8 Observation or Measure- ment from Outside

Applicable to all kind of structures, but necessary of excavation for hurried structure.

II Indirect Method

II-l Observation of Pipe Flow

Observe the pipe flow from man-holes; Applied for prelimi- nary investigation.

II-2 Measurement of Pipe Flow

Measure the pipe flow by using flowmeter at man-hole.

213
Investigation Method Feature and Applicability

II-3 Sounding by Rod Know roughly the location of blockade of small-diameter

pipe by insertion of rod.

II-4 Pumping Test between

Man-holes Know the inflow of groundwater by pumping out test of blocked section of pipe between man-holes.

II-5 Pumping Test at Joint

or Crack By using special device to block partially the water flow in the pipe at joint or crack, know the damage degree at joint or crack from the inflow of underground-water at this particular point.

II-6 Pore Water Test Applicable to pipe of existing over ground water table; Block the pipe by packer, pore water into this block the

section, measure the decrease of water table to know the damage.

II-7 Pneumatic Test and Smoke Test Send smoke into pipe from man-hole, know the damage of

pipe by observing the leakage of smoke; or send compressed

air into the blocked section of pipe to know the damage

from air-tightness.

II-8 Underground Rader Know the location of hurried pipe by underground rader.

II-9 Water-Quality Test Know the possibility of inflow of groundwater by measure- ment of chlorine ion.

214
Table 5.3 Repairing Method of Damaged Pipe and Man-hole

(a) Pipes Damage Form Method for Large-Diameter Pipe Method for Small- Diameter Pipe

1 Breakage of Pipe Concrete Coverage, Replacement Replacement

2 Crack of Pipe Resin Injection Replacement

3 Breakage of Joint Mortar Plug after V-form Cut Improvement of Water-tightness by Chemiceil Grouting

(Slight damage), Replacement

(Heavy damage)

4 Dislocation at Joints

Sealing, V-form Cut and Mortar Plug, Bandage against Water LeakeChemical Grouting using Packer, Replacement

5 Mixture of (T) ~ Concrete Lining, Replacement Double Tubing, Replacement

6 Vertical Snaking Improvement of the Lie by Special Chemical Grouting, Replacement Improvement of the Lie by

Special Chemical Grouting, Replacement

7 Lateral Snaking Replacement if impossible to repair by (D~(D

Replacement if impossible to

repair by (g)~(^

8 Uplift, severe Dislocation of

Pipe Axis

Replacement Replacement

(b) Manholes Damage Form Repairing Method

Dislocation of Prefabricated Wall Joints Replacement of Prefabricated Elements

Crack of Prefabricated Wail Chemical Grouting (Slight Damage)

Replacement of Damaged Elements (Heavy Damage)

Crack of Cast-in-place Wail Resin Injection or Chemical Grouting (Slight Damage)

Reconstruction (Heavy Damage)

Crack of Invert Resin Injection (Slight Damage)

Reconstruction (Heavy Damage)

215
Table 5.4 Actual Example of Repairing Method (Nihonkai-Chubu Earthquake)

Damaged Portion Damage Form Repairing Method

Break Replacement

Leakage Replacement

Pipes Lateral More than 1/3 of Thickness of Pipe Replacement

Dislocation

Less than 1/3 of Thickness of Pipe V-form Cut and Corkage

Vertical Dislocation Replacement

Cast-in-place Wall of Man-hole

Break Reconstruction Crack V-form Cut and Mortar Corkage Movement Improvement of Invert

Prefabricated Wall of Man-hole

Break Replacement

Leakage V-form Cut and Mortar Corkage

Dislocation Severe Replacement with Excavation

Slight Improvement with Mortar

Invert of Man-hole

Break Reconstruction Leakage Improvement of Invert Crack V-form Cut and Mortar Corkage

Service Pipe Break Replacement

Projection Improvement of

216
M

(Magnitude

in

Richter

Scale)

8.0 7.9h

(The Tokachi-oki Earthquake)

7.7h

I (The Nihonkai-chijbu Earthquake)

7.0-

(The Niigata Earthquake) O

(The IVlivagiken-oki Earthquake)

ry IVIarginal line of lique faction given by \ Kuribayashi and Tatsuoka

Monitary amount of damage by unit length o 100,000 Yen/m o 10,000 O 1,000 o 100 o 10

• No damage

50 100 200

Distance from Epicenter (km)

500

Fig. 5.1 Ground Motion to Cause Damage of Sewage System

217
(b) Sand sedimentation in a manhole

Fig. 5.2 Damage of Sewage Facility due to the Ground Liquefaction

218
219
220
221
EFFECTS OF VERTICAL CONSTRUCTION JOINTS ON THE DYNAMIC RESPONSE OF ARCH DAMS

BY

H. L Boggs^ and J„ R. Mays^

ABSTRACT

A finite element structural analysis of

a double curvature thin arch dam which models the nonlinear behavior of the vertical contraction joints is presented. This static analysis

represents a first step in a project at the Engineering and Research Center of the United States Bureau of Reclamation

to examine the nonlinear behavior of a dam, foundation, and reservoir system

subjected to seismic loading.

The "Curved Isoparametric Interface Surface Element," described in a techpicjl note by Buragohain and ShahLlJ was coded and installed in the 1978 version of ADINAL2J^ jhe element

proved to be effective when a sufficiently fine mesh is used and when

an appropriate integration order is used

in the isoparametric element.

The static loading chosen to effect the opening of the vertical contraction

joints consisted of gravity plus

reservoir at approximately half height

of dam plus a temperature increase of 24°F on the section of the dam above water level. To isolate the nonlinear

effect of the joint opening, identical

loads were applied to an identical model

in which the vertical joints were allowed to carry tension.

The results of the two analyses are presented in the form of colored contour

maps of arch and cantilever stresses on both the upstream and downstream faces

of the dam. The inclusion of the nonlinear joint is clearly effective in reducing the calculated tensile arch stresses on the upstream face. The load redistribution from the more flexible to the stiffer sections is more subtle but graphically shown as stresses

in both the horizontal and vertical

di recti ons . KEYWORDS: Arch Dam; Contraction Joints; Finite Element; Nonlinear Analyses

1 . I nt roduct i on

1 . 1 General Concrete arch dams during design are idealized as homogenous, isotropic

and monolithic concrete structures.

These assumptions are varied during

construction but become actual after completion of construction. That is, during construction, concrete is placed as a heterogenous mixture of water, cement and aggregate which

soon hardens into a homogenous mass.

The material properties, not exactly

equal in each direction, are sufficiently close to isotrophy. Current construction practices of horizontal lifts in vertical blocks, which when artificially and uniformally cooled to dissipate the heat of hydration, cause the blocks

to shrink or contract thus separating

the adjacent blocks at the formed joints. These contraction joints,

which separate up to 3 mm, are soon filled with cement grout restoring

the dam to a monolithic arch structure as was assumed in design.

It is this last assumption during the life time of operation that may occassi onal ly be violated.

^United States Bureau of Reclamation,

Denver, Colorado 80225

2un iversity of Colorado at Denver, Colorado 80202 *Refers to references at end of manuscript

222
1 . 2 Loads

Arch dams are designed to support a constant dead load, a fluctuating

reservoir level, the annual concrete temperature variation from ambient

air and reservoir temperature, ice and silt when appropriate. Except

for some seasonal loads, most of these load combinations produce

compression in the arches. Load combinations tending to produce tension in the arches occur when the cold winter temperatures act in concert with a full reservoir and late in the summer when the reservoir

down from need and the temperture

of the concrete is high. In this case the confining water pressure is insufficient to overcome expansion

upstream from the increased concrete temperatures

.

The potentially more severe but less frequent load is the earthquake which combines structural oscillations with

the static loads occurring at that time. Generally the MCE (Maximum

Credible Earthquake) events are sufficiently large to cause tension

in both the horizontal and vertical directions on each face over the duration of the earthquake. The low reservoi r/ summer temperature load combination was selected to exercise

the contraction joint finite element

prior to analyzing the arch dam during an earthquake.

1 . 3 Analyses

Present analyses, using high speed

large capacity digital computers, use linear elastic finite element modules. Throughout the duration,

large tensions that develop are assumed not to cause structural distress to the dam. However, i the prototype, arch tension does not develop because of the formed cracks (contraction joints) and inherent weakness of the cement grout. At that instant in time, no horizontal tension exists, arch stiffness is reduced, and the loads are redistributed to the adjacent elements. This presumptive nonlinear

action needs to be qualitatively and quantitatively determined. During an earthquake, relaxation of .arch stiffness is exfrected to produce

additional structural damping and

mitigate the potential structural distress from the MCE.

At this time in the Bureau of Reclamation, a finite element has been developed to simulate the nonlinear no tension contraction

joints. Initial analyses to confirm

load redistribution were made with

the aforementioned low reservoir and summer temperature load combination

applied to two identical arch dams differing only in their ability to transmit tension across certain contraction joints. Following are descriptions of the finite element formulation, structural analyses,

results and evaluation.

2. FINITE ELEMENT PROGRAM

2 . 1 General

The 1978 version of ADINA (Automatic

Dynamic Incremental Nonlinear

Analysis )1-2J was selected as the finite element code to be used in this analysis. Modifications were made in the code to allow the addition of the interface element

used to model the vertical contraction joints in the arch dam.

A full Newton iteration procedure was added to complement the existing

modified Newton procedure for maintaining control of equilibrium.

2 . 2 Interface Surface Element

The "Curved Isoparametric Interface Surface Element," described by D. N. Buragohain and V. L. ShahL^J in a technical note was coded and installed in ADINA. This element

allows modeling of the contact surface between two adjacent blocks

of a structure. The number of nodes

on each surface of this element of zero thickness may vary from four to eight nodes. Two material models

have been used. The first, which assumes both tension and compression capability, essentially

glues the two construction blocks together. The second material models

the compression but no tension property, characteristic of blocks which are in contact with each other.

The stiffness of the joint element in the direction normal to the contact surfaces is controlled by the input material parameter Kp. So that significant overlapping does not

223
occur, Kn 1s chosen such that the stiffness of interface element is several orders of magnitude stiffen

than the material in the adjacent

blocks. Shear stiffness is controlled by a different material parameter Kg. The integration order

and thus the number of integration

points used in the normal isoparametric formulation of the element stiffness matrix is controlled by the input parameter

INT. The strain in the direction normal to the surface element as determined by the most recent displacement

configuration of the nodes is examined at each integration point.

If the strain indicates compression

at this integration point, then full values of both Kp and Kg are used to evaluate the contribution to the element stiffness matrix from the integration point. If the sense of the strain indicates tension at the point, then both Kp and Kg are taken

as zero. The element stiffness of the interface element is thus determi ned

.

2 .

3

Verification of the Element

The verification and characteristics

of the interface element were determined by modeling two cubes of material, one placed on the top of the other. The finite element model consisted of three-dimensional

elements to model the cubes with the interface element used to model the contact region. A load was used such that theoretical results at the contact surface were known. Some sensitivity of integration order and relative values of stiffness were noted. Good results were achieved if the integration order of two or three

and relative stiffness of no more

than two orders of magnitude were

used .

3 . FINITE ELEMENT MODEL

3 . 1 Dam and Foundation Morrow Point Dam was selected as the analytical model. This 465-foot-high

double curvature arch dam has a minimum thickness of 12 feet, a maximum thickness of 75 feet, and a crest length of 720 feet. Twenty-noded three-dimensional finite elements were used to model the dam.

Two elements through the thickness of the dam resulted in 128 elements.

The modulus of elasticity in the dam

was taken as 3x10® Ib/in^. The foundation was extended approximately

the height of the dam in all appropriate directions.

Th r ee -hu nd red twelve eight-noded

elements having a deformation modulus

of 2.5x10® Ib/in^ were used in the foundation. Morrow Point Dam, nearly

symmetrical as constructed, was made symmetrical in this analytical

exercise to reduce computational time

and number of elements.

3 . 2 Vertical Construction Joints

Many vertical contraction joints

exist in Morrow Point Dam. The finite element model, however, used only five vertical joints, placed

symmetrically on the structure. A total of fifty 16-noded interface surface elements were used. Again,

two elements were used through the thickness of the dam.

3 . 3 Complete Model PATRAN^^^, a commercially available computer code, was used to develop

the geometry and finite element mesh

for the complete model as shown on figure 1*. The resulting model consisted of 3,570 equations with a mean half band width of 202.

3 . 4 Loads

The final loads consisted of gravity, water elevation to approximately

mid-height, and a positive

temperature of 24°F throughout that section of the dam above the water

level. This combination of loads was chosen to exercise the opening of the vertical contraction joints. The element birth option of ADINA was used to bring the five vertical contraction joints into existence

after all loads were applied. This approximates reasonably well the order of block building and grouting

of the contraction joints. The sequence of loading steps are: full gravity applied; vertical construction joints are born (grouted); waterload to 240-foot

level applied; temperature of the dam above 240-foot elevation increases in 3°F increments until 24°F increase is *See figures at end of manuscript

224
reached. The temperature load is uniform between faces and abutments.

4. PRESENTATION OF RESULTS

PATRAN^^^^ was used both as a preprocessor and a postprocessor for ADINA. PATRAN was used to prepare

contours of equal stresses drawn separately on the upstream and downstream face of Morrow Point Dam. The two types of stress quantities drawn

were the arch and the cantilever

stresses on the faces. Both the arch

and cantilever stresses are normal stresses lying in the plane tangent to the face of the dam at the point in question. The arch stress is in a horizontal plane while the cantilever

stress is in a vertical radial plane. A complete translation code was written as an interface between ADINA and PATRAN.

This translator resolved the global stresses into the arch and cantilever

directions. The accumulative effect of this incremental loading on a linear elastic structure is graphically shown

on figures 2 and 3 in the form of arch

and cantilever stresses on the upstream

face. Arch dam stresses from construction by vertical blocks are shown on figures 2a and 3a. Note on figure 3a the development of vertical tension near the crown cantilever where the downstream

overhang is the greatest and simultaneously producing compression in the lower

half where the undercutting is also greatest. This action with horizontal stiffness produces small horizontal

stresses shown on figure 2a. The addition of water to mid-height alters the cantilever stresses little as shown on figure 3 but significantly increases the arch stresses as shown on figure 2. The incremental addition of temperature

shows the rate of tension developed and magnitude of load necessary to produce

sufficient arch tension for this analytical exercise. The significant

change in arch and cantilever stresses

are contained to the upper half of the dam where the loads are applied as seen

on figures 2c-2f and 3c-3f. The effect

on arch and cantilever stress from contraction joint opening, is shown on figures 4 and 5. To emphasize the dramatic load shift, the upstream and downstream arch and cantilever stresses

are subdivided in the tension zones as shown on figures 6 and 7.

Eval uati ons

The load combination of low reservoir,

summer temperatures and gravity produces

a stress reversal to the normal reservo ir/winter load combination that

is of more concern to the designer.

This latter load combination greatly

influences the structural shape so as to cause compression in the arches, on both faces, base to crest, and between abutments. Similarly, most of the cantilevers are totally in compression,

except small tensions along the abutment upstream face and the upper central section on the downstream face.

On figure 4a note the large area in tension caused by temperature expansion

which completely disappears, figure 4b,

if the interface element modeling the contraction joint is allowed to open rather than carry tension.

In the total elastic analysis,

temperature expansion of the arches,

causing tension on the upstream face,

naturally causes compression on the downstream face. Relaxation of the arch stiffness via the contraction joint

opening also tends to dissipate the downstream compression as shown in comparing figures 4c and 4d. Temperature expansion also causes vertical tension in the cantilevers as

is seen on figure 5a. Concurrently,

the arch stiffness relaxation also reduces the cantilever restraint which simultaneously eliminates cantilever

tension as is seen on figure 5b and compression on the downstream face,

figures 5c and 5d. The total dissipation of tension does

not occur with the limited number of contraction joint elements as in this model as can be seen by recontouring the upstream stresses of the arches on figures 6a and 6b, and cantilevers,

figures 7a and 7b. These small arch tensions would further dissipate by adding contraction joint elements in between the present configuration. All cantilever stresses in tension may not completely dissipate except in the upper

part where no arch tension or compression exists. In other areas, the existence of arch compression signifies

arch stiffness and does not relieve all cantilever tension.

225
Conci us ions Linear elastic analyses provide some measure of structural distress during

load combinations producing tensile stresses such as during the low reservoir, summer temperature, and gravity load combination or as with an earthquake event. Reducing the element stiffness transfers sufficient load to adjacent structural components to mitigate or eliminate all tension. Recommendat i on Based on the load transfer following the reduced stiffness in elements not capable of tension, this finite element process is applicable to solving the in-elastic problem of arch dam structural response during an extreme seismic event. REFERENCES

[1] Buragohain, D. N. and V. L. Shah, "Curved Isoparametri c Interface Surface Element," Journal of the Structural Division, ASCE, Vol. 104, No. STl, January 1978, pp. 205-209.

[2] "ADINA - A Finite Element Program

for Automatic Dynamic Incremental Nonlinear Analysis," Report 82448-1 (Revised Dec. 1978) , ADINA Engineering, December 1978.

[3] "PATRAN - G Users Guide", PDA Engineering, Santa Ana, CA.

226
X

cn

lij zM

ll.

(9M

U.

227
FIG.

2

ARCH

STRESSES

(PSI)

-

UPSTREAM

FACE

228
FIG.

3

CANTILEVER

STRESSES

(PSI)

-

UPSTREAM

FACE

229
FIG.

4

ARCH

STRESSES

(PSI)

230
FIG. 5

CANTILEVER

STRESSES

(PSD

231
FIG.

232
FIG. 7

CANTILEVER

STRESSES

(PSI)

233
DECISION ANALYSIS FOR EMBANKMENT DAMS

UNDER SEISMIC LOAD

BY

T. A. Luebkel

ABSTRACT

Decision analysis may be defined as a process of systematic evaluation of alternative solutions to a problem where

the decision is made under uncertainty.

This method of analysis readily applies

to seismic design and analysis of embankment dams because both the seismic

event and response of the embankment are uncertain. The decision analysis does

not replace the design process, but instead presents the risk and consequences of dam failure due to seismic loading for the various alternatives

examined. The analysis is made up of two parts: (1) the hazard assessment

and (2) the risk cost analysis of alter- natives.

This paper presents a framework for decision analysis that may be applied to embankment dams in seismic areas. Simplified methods are described to determine the seismic risk of dam failure. Considerations in the deter- mination of recurrence intervals for seismic events are briefly discussed.

However, the paper concentrates on the application of the recurrence estimates

in evaluating risk, and on the deter- mination of potential for loss of life

and adverse economic consequences. An example problem is provided to illustrate the use of decision analysis

and the presentation of data for decis i onmaki ng. KEYWORDS: Dams; Hazard; Risk; Seismic

1 . INTRODUCTION Decisionmaking in situations where the outcome is uncertain is a difficult task

to verify by quantitative procedures.

Such decisions must be made for selecting appropriate dam designs in seismically active areas of the world.

Our current understanding of the pro- cesses by which seismic activity occurs

is somewhat limited, as is our ability

to accurately predict the timing and magnitude of the next seismic event

at a given source. Moreover, our ability to confidently predict the response

of dam structures to seismic events is

derived from empirical data. Although

earthquakes occur on a relatively fre- quent basis around the world, our knowledge of how dams react to seismic events is generated from a relatively

small data base. Given these significant uncertainties, dams located in seismic regions must be designed such

that they will not represent a significant hazard to the public, while main- taining fiscal responsibility.

This paper presents a framework for decision analysis which provides a means

for displaying uncertainty in a manner

which allows an informed decision. The decision analysis does not provide answers, only information to be con- sidered. The procedures are most useful when evaluating complicated problems where solutions are not obvious and justification for a particular course of action is required. A brief description

of the decision analysis framework is provided, followed by an example problem

to illustrate its use. The scope of the example problem is limited to seismic- related hazards and risks associated

with a particular embankment dam.

2. DECISION ANALYSIS FRAMEWORK

The decision analysis framework consists

of an organized investigation and pre- sentation of alternatives for accepting

or reducing potentially adverse con- sequences and/or risk associated with system failure or emergency operation.

A general framework (reference 1) is depicted on Fig. 1. The framework can be condensed and more easily understood by first identifying

the two major phases in the formulation

of information in this analysis:

(1) hazard assessment and (2) risk cost analysis of alternatives.

The hazard assessment is performed to evaluate the consequences of dam failure

and/or emergency operation (e.g., loss

1 Bureau of Reclamation, PO Box 25007, Denver, Colorado 80225

234
of life, economic loss, environmental

impact, etc.) for a given dam safety

problem. This information is used to establish the minimum acceptable level

of protection required, if the dam safety problem involves an existing

structure. New structures generally

will be designed to accommodate the maximum loading condition without

catastrophic failure [e.g., PMF (probable maximum flood), MCE (maximum

credible earthquake)]. The hazard assessment may often be effective in identifying alternatives. If an appropriate solution to the problem is identified and adequately justified at this stage, then a risk cost estimate

for the existing condition and for various alternatives would not be requ i red

.

Once the minimum acceptable level of loading has been identified (based on appropriate input from the decisionmaker), alternative courses of action

are developed and may be compared using

risk cost analysis of the adverse con- sequences (identified in the hazard assessment), the likelihood of adverse

consequences, and the cost and benefits

of corrective action. Depend-ing on the results of the hazard assessment, the

risk cost analysis may also be performed

to determine whether a greater level of protection is economically justifiable.

The results of the hazard assessment and

any required risk cost analysis form the basis for presenting information to the decisionmaker so that the s i te- spec i f i

c

conditions and options are well understood.

The decision analysis framework may be broken down into a stepwise process,

with presentation of results for management decisions at appropriate points

during the process. The framework is illustrated by an example problem in the following section of this paper.

3. EXAMPLE

3 . 1 Problem Definition

The example problem provided in this

paper is partially based on data obtained from an existing dam. The analyses and judgments presented in this

paper are solely for the purposes of illustrating the application of decision

analysis for embankment dams under seismic load, and do not reflect final ana- lytic results nor final judgments made.

The problem analyzed in this paper is the dam's ability to withstand potential seismic loading. If the dam is found to be subject to failure under seismic

loading, alternative courses of action

must be examined.

3.2. Pertinent Site Conditions

A description of pertinent site con- ditions is provided to summarize infor- mation which has a bearing on the analytic results. The example dam is located in the western United States.

The reservoir serves as a link in crossbasin transfer of water along the south

slope of the Uintah Mountains between

two existing reservoirs. Dam construction was completed in 1975. Initial filling has taken pi ace. Major features include the dam embankment, spillway, and outlet works. The embankment is a zoned earthfill structure with a maximum height above streambed of approximately 130 feet and

a crest length of 1,600 feet. The spillway is a box-type conduit structure

with an uncontrolled ogee crest. Discharge capacity of the spillway at the maximum water surface is 850 ft^/s.

The outlet works is a cut-and-cover con- duit structure. A steel pipe branch

from the downstream control structure connects the outlet works to a pipeline

feeding a nearby reservoir. The outlet works is also capable of discharging

through a stilling basin into a creek. Discharge capacity of the outlet works

at the top of active conservation is 5,430 ft3/s.

Operating and failure discharges are released into the creek, which drops

1,400 feet over 21 river miles before

its junction with a second creek, which descends 200 feet over 4 river miles to the main river. From this confluence,

the river drops another 400 feet before entering a nearby reservoir 15 river miles downstream.

A review of construction records indi- cated that the embankment is a densely

compacted fill with a clayey core and granular shell materials. A cutoff

trench was provided to bedrock.

However, geologic data indicated that alluvial deposits up to 20 feet in thickness remain under the shells of the embankment. The alluvial deposits consist of heterogeneous layers of silty to clayey fine sand, sand and gravel, organic silty sand, and peat. The finer grained materials (lacustrine deposits)

were deposited in oxbows, ponds, and

235
marshes and are laterally discontinuous.

The small amount of SPT (standard

penetration test) results (all from preconstruction drilling) reported

N values of 4 to 18 blows per foot. The upper portion of the foundation material

was largely removed (including the peat

and loose fine-grained material).

However, some materials with low SPT blow counts still remain. The geologic

sections indicated a general discontinuity of the lacustrine deposits and dominance of the gravelly streamfill deposits on some sections, but con- tinuity of several hundred feet of the lacustrine deposits were reflected on other secti ons

.

3 . 3 Potential Loading Conditions

Seismotectonic considerations for the example dam are summarized in Table 1. The table indicates seven distinct

sources for MCE's between magnitude 6.5

and 7.5 within 90 km of the dam. Fig. 2 indicates the relationship of bedrock

peak horizontal ground acceleration versus horizontal distance from source of energy release adapted from reference 2, The earthquake magnitude ranges from Table 1 are plotted onto Fig. 2, indi- cating that mean peak horizontal bedrock accelerations of between 0.10 and approximately 0.55 may occur at the damsite. Since it is known that the dam is in direct contact with bedrock at the cutoff trench and the alluvial fill under the shells is relatively shallow,

the bedrock motion is not expected to attenuate prior to its loading the base

of the dam.

It is important to note that the final seismotectonic considerations reported

that surface faulting at the damsite and reservoir-induced seismicity were not considered to be potential hazards. Reservoir seiche waves and landslideinduced water waves were not felt to be significant issues.

sources are plotted on the figure based

on epicentral distance for the general

range of earthquake magnitudes that fall above the line (capable of producing

liquefaction). Note that each of the identified seismic sources are empirically capable of producing liquefaction.

It is assumed, for purposes of illustration in this paper, that more detailed

analytic studies have been performed and support the findings of the empirical assessment. It is also assumed that liquefaction of the alluvial materials

beneath the embankment shells will result in dam failure. The next step is to determine system response in terms of downstream flooding due to dam failure.

The active conservation pool for the example reservoir is expected to remain relatively constant, within 4 feet of top of active conservation at all times.

Dam breach studies indicate that peak

discharge resulting from dam failure

will be 300,000 to 400,000 ft^/s. This discharge will be reached within

30 minutes of the triggering seismic

event. Inundation maps are prepared to depict the downstream areas affected by dam failure. Expected depth and velocity of the failure flood are considered.

It is determined that the failure flood inflow to the downstream reservoir, located approximately 40 miles downstream of the example dam, can be compl etely stored.

3 . 5 Hazard Assessment

The hazard assessment is performed to evaluate the consequences of dam failure

in terms of loss of life, economic

impact, environmental impact, and other

losses. Buildings within the potential dam failure inundation area between the example dam and the downstream reservoir

have been identified as follows;

3.4. System Response

An empirical assessment of the capability of the MCE's provided in Table 1 to produce liquefaction in the foundation

soils (assuming they are liquefiable)

can be made by plotting the MCE's for each source onto Fig. 3. The dashed

lines on Fig. 3 represent the threshold

boundary for earthquakes with the potential for producing liquefaction in liquefiable materials based on historical data. Although specific values are

not available, it is implied that the lower boundary represents materials very susceptible to liquefaction. The fault

Building Locati on

1 residence 8 river miles below

dam

5 residences and 12 to 13 river miles

2 commercial below dam. near buildings Highway 40

201 residences 20 to 40 river miles below dam

The total permanent resident population

in the dam failure inundation areas has been estimated at around 80 to 100

236
people depending on the time of year. A few of the residences are vacation-type

dwellings. However, during the late spring and summer, these units would probably be occupied and so a permanent

residence population at risk of 100 was assumed under worst case conditions.

During the late spring and summer

months, the local creeks and rivers pro- vide recreational fishing and camping.

There are, however, no established

(State, Federal, private, etc.) camp- grounds along the 40-mile stream course

between the existing dam and the down- stream reservoi r. Based on cursory field reconnaissance,

it has been estimated that during the late spring and summer months the number

of transitory people in the dam failure

flood inundation area would range from

10 to 50 people for fishing and camping,

and 5 to 10 people for other miscellaneous purposes during the daytime. At night it was estimated that there would

be 5 to 20 campers in the flood plain

area and 0 to 5 people traveling through

the area for other miscellaneous

reasons. It may be assumed that the transitory population would be more or less uniformly distributed along the flood reach, except for the undeveloped

and relatively inaccessible reach 13 to

20 river miles below the dam, and that

the spatial distribution of the resident

population at risk would be according to structure locations, as presented in Table 2. The table indicates that the population at risk from dam failure is relatively small. However, the population is spread out over a large

geographic area. Further, the tran- sitory population cannot be easily con- tacted (e.g., by telephone) in the event

of a failure. Table 3 provides estimates of the time required for dam failure floodflow to reach each of the downstream areas.

The estimate of warning time available

for the population at risk is based on the detection and warning capabilities

of the system. A dam tender will not be onsite but the system will be operated

and monitored (reservoir level and downstream discharge) by remote control on a 24-hour basis. The times at which

people within the various river reaches

are likely to be warned, both by

telephone and in person, are shown in Table 4. The table also shows the number of residents and transitory visitors in each river reach, and when the failure flood arrives at each section.

It was assumed that 30 minutes would be

required to detect the drop in reservoir

level resulting from a sudden dam failure. At that point, monitoring

officials would notify local authorities

and begin telephoning residents in the upper reaches of the river.

It is estimated that 15 minutes would be required to notify local officials and begin warning residents by telephone of the probable dam failure flood. As shown jn Table 4, it is estimated that neither telephone nor face-to-face

warning can reach the population at risk before the failure flood arrived, except

for river sections E, F, G, and H, where warnings can reach the residents shortly before the failure flood.

The probable loss of life resulting from downstream flooding can be estimated

using the procedures described in reference 1. The reference includes an analysis of 24 historic dam failures and flash floods. These case studies have

been evaluated to determine the rela- tionship between the number of persons

at risk, the amount of warning time available, and the number of fatalities

resul ting.

To estimate the loss of life, the popu- lation at risk is first divided into two groups: those people receiving adequate

warning time to evacuate the flood area

and inadequate warning time to evacuate

the flood area. Based on empirical data from flash floods and dam failures, this break appears to be between those per- sons receiving more than and those receiving less than 90 minutes of warning. Separate predictive equations,

shown below, may then be used to provide

an initial (not site-specific) estimate

of loss of life for these two sub- popul ati ons . Loss of life, given 90 minutes or more warning = 0.0002 x (population

at risk) Loss of life, given less than 90 minutes warning = (population at

r i s k ) 0 • ® These equations predict the loss of life

that is likely to occur, based on his- torical averages, for a given population

size and either adequate or insufficient

warning time. Note that for the situation with less than 90 minutes

warning the equation projects less than

100 percent fatalities. This is because historically many individuals at risk receive some minimal warning, while

other persons detect the flooding on their own, and evacuate, while still

237
others are able to survive the

f 1 oodwaters

.

These equations can be used to make "baseline" projections of loss of life.

The estimates are usually adjusted up or down if the particular setting poses unusually threatening or unusually

benign conditions, compared to the historical cases of dam failure and flash flooding upon which the equations

are based. The equations provide esti- mates of the most likely outcomes, given

the population at risk and warning time. Many other factors (e.g., the ease of evacuation from the flood plain, lethality of the flood in terms of depth,

velocity, and ability to carry debris,

etc.) can affect the actual outcome, and these should be considered. The reliability of prediction is much greater for those cases involving ade- quate warning.

The equations used to predict loss of life require estimates of population at risk for each flood and the estimated

warning time available for each population. This information is shown in Table 5. As indicated in Table 5, a sudden dam failure is likely to place up to approximately 158 persons at risk without adequate warning. However, this condition will occur only during the summer months. Fewer residents and far fewer visitors are likely to be in the area during the spring, fall, and winter. The following table shows the baseline estimated loss of life for the maximum likely population at risk. Baseline Loss of Life Estimates For Sudden Dam Failure (Summer Season)

Population

at Risk

Potenti al Loss of

Li fe

Insuff i ci ent warning

( <90 minutes)

114-158 17-21

Baseline estimated total loss of life 17-21

In instances of sudden dam failure, where some populations may receive no warning of a flood, and where depths and velocities may be considerable, it is

appropriate to revise the baseline

fatality estimates upward. One of the worst cases of dam failure flooding,

with a population at risk of approximately this size and virtually no warning

(Laurel Run, PA, 1977), produced

approximately 30 percent fatalities. In the present case, approximately 48 to

74 persons would receive no official warning (reaches A through E) while approximate! y 66 to 84 persons would receive some advance warning. Applying

the 30 percent fatality rate only to those receiving no warning produces a revised fatality estimate of 26 to 36 (26 = (48x0.3) + 660-6; 35 = (74x0.3) + (84)0.6)^

The prior estimates of probable loss of life are based on existing conditions.

Realistically, over the next few decades, it is anticipated that local use, recreational usage, and population

development between the example dam and the downstream reservoir will probably

increase. As a rough quantitative indi- cation of probable future changes, it is estimated that the population at risk

will probably increase by less than

50 percent during the next two decades.

(The past two decades has seen a change

in permanent residents of about one-thi rd)

.

In addition to the estimated loss of life, the magnitude of economic impact

resulting from dam failure is on the order of $35 million to $110 million

depending on the availability of water

when the failure occurs (e.g., failure

during a relatively dry period will have

a greater economic impact than failure

in a wet period). The estimated impacts

can be divided into the following

categories:

Loss of irrigation benefits - $8 to

$ 8

1

million

Cost to repair the dam - $15 million Downstream property damage - $10 million

Loss of recreation benefits - $2 to

$ 3 million

Environmental impacts and other losses were determined to be insignificant for

thi s parti cul ar case.

In evaluating the adverse consequences

due to dam failure, the decisionmakers

first seek to determine the minimum

level of protection required. If there

is no significant incremental potential

for loss of life resulting from dam failure (or other noneconomic factor),

the analysis may continue by evaluating

the likelihood of events to produce the

238
adverse impacts to determine an optimum

level of expenditure to reduce the risks. However, if there is a significant incremental potential for loss of life, the maximum loading condition is chosen for design purposes. For earthquake loading, the maximum loading con- dition may be the MCE.

For the example dam, it has been shown

that there is a significant incremental

potential for loss of life in the event

of sudden dam failure. Therefore,

unless the population can be relocated,

it was determined that the dam must be designed to withstand the MCE, regardless of the probability of occurrence of that event. The risk cost analysis of alternative actions would

only consider alternative approaches to designing for the MCE

.

For the purpose of illustration in this

paper, it is assumed that the population

can be relocated and the flood plain

managed so that the potential for loss

of life from dam failure is extremely

low. It is assumed that the relocation

can be accomplished for $10 million.

Unless a means of precluding liquefaction can be found for less than

$10 million, this alternative appears to

be economically justified without con- sideration for the potential economic

losses associated with project benefits

and dam repair following dam failure. Further analysis of these issues is pre- sented next.

3 . 6 Risk Cost Analysis

A risk cost analysis is performed to estimate the risk associated with the existing conditions and the risk reduction accomplished by alternative courses

of action. This process facilitates

comparison of proposed actions and leads

to a good understanding of the nature of the problem.

Quantitative risk assessment requires

consideration of the likelihood of dam failure and the adverse consequences

resulting from dam failure. An eval- uation of the probability of the seismic sources to produce liquefaction

at the damsite can be performed based on the empirical assessment presented in a previous subsection of this paper. Table 1 presents a range of estimated

recurrence intervals for each earthquake

magnitude range from each source. Note

that the four segments of the Wastach

fault are treated as independent sources

in the probability estimate. The recurrence (average return period)

ranges reflect a range of uncertainty

for the corresponding range of earth- quake magnitudes and do not reflect the assignment of a specific recurrence

interval to a specific earthquake magnitude. Data in the last column includes

some judgment of the time interval to the next seismic event based on the time elapsed since the last event.

From the sei smotecton i c information and the assumptions previously stated, a total annual probability of liquefaction

due to the occurrence of seismic events

is estimated using the follow i'ng equation:

PT = 1 - [(1-Pi)(1-P2)(1-P3)...(1-Pn)]

where Py is the total annual probability

and Pi, P2, and P^ are the individual

annual probabilities of the seismic

events large enough to produce liquefaction at the damsite for each fault source (which equal the inverse of the recurrence i nterval ) . For earthquake sources which have two distinct ranges of recurrence (e.g., the random earthquake in this problem), the following equation was used to estimate

the individual annual probability from

the fault source if the two earthquake

ranges have a different capability of producing liquefaction at the damsite:

Pn = C (Pa - Pb) X PdfiI + [Pb X PdF2^ where P^j is the total individual annual probability of liquefaction for the fault source. Pa and Pb are the annual probabilities for the two distinct

ranges of recurrence, and Pdfi PqPp are the different probabilities

that the seismic source is capable of producing liquefaction, if the deposits

are suscepti bl e. The interpretation of the recurrence relationship for the random local vicinity earthquake is somewhat complicated because no specific epicentral

distance is assigned to the earthquake

magnitude; it simply represents a "near field" event. For this range of earth- quakes, near field represents approximately from 2 to 20 km, with a worst

case range of 2 to 5 km as plotted on Figs. 2 and 3

.

Pop, has been assigned a value of 1.0 since the upper magnitude range con- sistently falls above the liquefaction

threshold. Pqp

2 has been assigned a value of 0.5 to'^reflect that the lower magnitude range generally plots below

239
the threshold of liquefaction. An engineering adjustment has also been made

for the Wasatch fault sources at 68 km and 90 km to reflect the fact that much

of the earthquake range falls below the threshold. A value of 0.75 and 0.50 has been assigned to the respective sources

to obtain the probability of liquefaction. For all other seismic sources

under consideration, it was assumed that liquefaction would occur (probability of 1.0) if the seismic event took place.

A separate calculation is performed for each extreme of the recurrence range (where applicable) so that the total estimated probability of liquefaction is also a range. The calculations are per- formed as follows: Adjustments to obtain probability of seismic events capable of producin g

1 iquefaction

1. For random earthquake (least remote side of range):

Pa = 10-2 p |3 = 10-3 = 0.5 PdF2 " 1-0 Therefore, Pfj = [(0.01 - 0.001) x 0.5] + [0.001 X 1.0] = 5.5 X 10-3

For random earthquake (most remote

side of range) The Pfi = 10-3 X 1.0 = 10~3

2. For the Wasatch fault earthquakes

that require adjustment:

Bounti f ul Segment

:

Pn = 2 X 10-3 X 0.5 = 1 X 10-3 (least remote)

= 10-3 X 0,5 = 5 X 10-^

(most remote) Corner Canyon Segment:

= 10-3 X 0.75 = 7.5 x 10-^ (least remote)

= 5 X 10-^ X 0.75 = 3.75 x 10-^

(most remote)

Nephi Segment:

Pn = 2 X 10-3 X 0.75 = 1.5 x 10"3 (least remote)

= 4 X 10-4 X 0.75 = 3 x 10'4

(most remote)

Calculations of total probability of

1

i quefacti on Pj = 1 - [( 1 - 0,001 ) ( 1 - 0.00075 ) (1 - 0.002) (1 - 0.0015)

(1 - 0.0003) (1 - 0.0003)

(1 - 0.0055)] = 0.0113 for the least remote side of range

Pj = 1 - [(1 - 0.005)

(1 - 0,00375) (1 - 0.001)

(1 - 0.0003) (1 - 0.0001)

(1 - 0 . 0001 ) (1 - 0 . 001 )]

= 0.0034 for the most remote

side of range

Based on the preceding calculations, the extreme worst case estimate for the annual probability of seismic events

with the capability of producing

liquefaction is on the order of from

1.1 X 10-2 to 3.4 X 10-3 which is from

one to two orders of magnitude greater

than the historical dam failure probability from all causes. It equates to a seismic event capable of producing an earthquake large enough to cause liquefaction in susceptible deposits

occurring once every 90 to 300 years. Assuming that property and population

relocation has occurred, the adverse consequences resulting from dam failure

due to liquefaction are estimated to result in economic losses from

$25 million to $100 million. Then the range of annual seismic risk costs asso- ciated with the existing structure is from $85,000 to $1,100,000.

3 . 7 Alternative Actions

Based on the information provided in previous sections of this paper, three courses of action may be considered:

A . No action.

B. Relocation of the population and flood plain restrictions.

C. Modifications to the dam to preclude dam failure due to

1

i quefacti on . D. Removal of the dam from service. Alternative A was eliminated from con- sideration because there is a potential

for failure due to loading conditions

less than the maximum (in this case the

MCE) which results in a significant

potential for loss of life. Alternative B eliminates the potential

for loss of life at a cost of

240
$10 million. Using the present discount

rate, the annual cost would be approximately $800,000.

Alternative C has been determined to cost $25 million for the least cost solution to the liquefaction problem.

Using the present discount rate, the

annual cost for modification of the dam would be approximately $2 million.

Alterpative D has been determined to cost approximately $12 million in average annual lost project benefits.

The results of the decision analysis are provided to the decisionmakers. Various conclusions may be drawn from the analysis for the example dam. A possible

decision may be that alternative B is selected while alternatives C and D are rejected due to their relatively high

annual cost in comparison with the annual risk cost.

4. SUMMARY AND CONCLUSIONS

This paper has shown that a decision

analysis framework may be useful in evaluating seismic risk associated with embankment dams. Analyses are provided

to determi ne:

• The loading conditions that cause

dam failure.

• The system response to the loading

conditions.

• The adverse consequences of dam failure in terms of loss of life

and economic impact.

• The probability (risk) of dam failure due to seismic loading.

Based on the analyses, alternative courses of action may be developed and com- pared to determine the most reasonable

solution to the problem.

5. REFERENCES

(1) Bureau of Reclamation, "Guidelines

to Decision Analysis," ACER Technical Memorandum No. 7, Denver, Colorado,

March 1986.

(2) Seed, H. B., and Idriss, I. M., "Ground Motions and Soil Liquefaction

During Earthquakes," Earthquake

Engineering Research Institute,

Berkeley, California, 1982.

241
Figure 1.

1

- General framework for decision analysis.

242
Table 1. - Magnitude and recurrence estimates for maximum credible

earthquakes of possible concern to the example dam

Se i smogen i

c

structures Magnitude

range

(Ms)

Fpi central distance

(km)

Range of average

return periods

(ka)*

Ag e of most recent

event

(ka)*

Range of est i mated

annual probabilities

of occurrence

Wasatch fault

Bri gham Ci ty- Bounti f ul Segment6. 5-7.

5

90 0. 5-1.0 0.3-0.

5

2x10-3 - 10-3

North Salt Lake City-Corner

Canyon Segment

6. 5-7.

5

68 2. 2-4.

6

>0.13 0

1 CO 1 cn O

A1 pi ne-Span i sh Fork Segment

6. 5-7.

5

48 1.7-2.

5

0. 3-1.0 2x10-3 - 10-3

Payson-Nephi

Segment

6. 5-7.

5

68 1 O

LD O 0. 3-0.

5

2x10-3 - 4x10-4 Strawberry fault 6. 0-7.0 10 3-10 >3 3x10-4 - 10-4 Stinking Springs

fault

6.0-6.

5

17 3-10? - 3x10-4 - 10-4

Random earthquake 1 in o Local vicinity

0.1-10? - 10-3

5. 0-6.0 Local vicinity0. 1-1.0? - 10-2 . 10-3

* ka = 1 ,000 years

.

243
O £

Figure 2. - Bedrock peak ground acceleration versus horizontal distance from source

of energy release.

244

(Iml)

HORIZONTAL

DISTANCE

FROM

SOURCE

OF

ENERGY

RELEASE

(kilomeler)

Adopted

from

Moon

Peak

Accelerolion

for

Rock

Site

(Seed

ond

Idnss

1982)
I

A NOLIOUEFACATIONIaTSITES SUSCEPTIBLE

LIQUEFACTION J TO LIQUEFACTION THRESHOLD BASED ON CHINESE DATA O LIQUEFACTION FROM JAPANESE EARTHQUAKES LOWER BOUND BELOW WHICH NO LIQUEFACTION SHOULD OCCUR

-

EE U G

s rs ”

o

r> E

( >.M i ° ^ J0*^ E E

c:v D M

zj Wasatch E

TTTT

k Strawbe: L4 i ie^ents

E E

N. 5 / i

i w E E

E.

doi F1 E

1 E

E f—

E E

E

E

"

TTTT E

I 5 10 20 SO 100 200 500 1000 EPICENTRAL DISTANCE (km) Figure 3. - Seismic potential at site vs. empirical liquefaction occurrence.

Table 2. - Total population at risk

River-mile location

below the example dam

Total resident population

at risk in flood area*

(day or night)

Total transitory

population at risk in flood area Dayt i me Ni ght

A . 0 to 5 miles 0 2-8 1-3

B. 5 to 10 miles 4 2-8 1-3

C . 10 to 15 miles 20 2-8 1-3

D . 15 to 20 miles 0 0-2 0

E . 20 to 25 miles 16 2-8 1-3

F . 25 to 30 miles 20 2-8 1-3

G. 30 t 0 35 miles 20 2-8 1-3

H . 35 to 40 miles 20 2-8 1-3 Total

s

100 15-60 5-25 *Assuming an average of four occupants per residential building.

NOTE: Totals may differ from column sums due to integer rounding of column values.

245
Table 3. - Flood travel times

River mile location

below the dam

Time (in hours) after initiation of dam failure when initial

dam failure flood waters

reach center of area

Ti me (in hours) after initiation of dam failure when peak dam failure flood waters reach center of area*

A 0 to 5 miles 0.1 0.8

B 5 t 0 10 miles 0.5 1.0

C . 10 t 0 15 miles 0,8 1.2

D . 15 to 20 miles 1.2 1.6

E . 20 to 25 miles 1.4 1,8

F . 25 to 30 miles 1.8 2.2

G , 30 to 35 miles 2.4 2.7

H . 35 t 0 40 miles 2.9 3.3

* Based on assumption of 0.75 hour requi red for the dam failure.

Tabl e 4. - FI ood travel and warning times (i n hours) : Sudden dam failure

R i ver mi 1 e 0 5 10 15 20 25 30 35 40 Study reach A B C D E F G H

Permanent pop- 0 4 20 0 16 20 20 20 ulation at risk Transitory pop- 2-8 2-8 2-8 0-2 2-8 2-8 2-8 2-8 ulation at risk Dam failure f 1 ood 0.1 0.5 0.8 1.2 1,4 1.8 2,4 2.9 arrival (10,000 ft3/s)

Telephone warning NF NF 0.75 NF 1.0- 1.0- 1.5- 1.5-

( begi nni ng- 1.25 1.5 1.5 2.0 2.0 end i ng

)

Face-to- face NF NF NF NF NF NF 1.0- 1.0- warning 2.0 2.0 Warni ng time _ 0 0 _ 0- 0.3- 0.4- 0.9-

(tel ephone)

prior to 10,000 ft3/s

0.4 0.8 0.9 1.4

( face-to-face) (-) (-) (-) (-:1 (-) (-) (0.4- (0.9-

1.4) 1.9)

to = Dam failure.

^0.5 Detection of event and decision to warn, initiation of telephone and physical

warnings.

NF = Not f easi bl e„

246
Table 5 Probable warning times for sudden dam failure flood hazard Resident popu- River mile location lation at risk below the example dam in flood areas

( day or ni ght)

Transitory popu- Hours of warning

lation at risk in prior to arrival

flood areas of dam failure flood

(day and night) (Telephone) (Face-to-face)

A. 0 to 5 miles 0 2-8 0 NF

B. 5 to 10 miles 4 2-8 0 NF

C. 10 to 15 miles 20 2-8 0 NF

D. 15 to 20 miles 0 0-2 0 NF

E. 20 to 25 miles 16 2-8 0-0.4 NF

F. 25 to 30 miles 20 2-8 0.3-0.

8

NF

G. 30 to 35 miles 20 2-8 0.4-0.

9

0.4-1.

4

H. 35 to 40 miles 20 2-8 0.9-1.

4

0.9-1.

9

Total s

100 1T=T8

NF = Not feasible.

247
A NEW CRITERION FOR ASSESSING LIQUEFACTION POTENTIAL

BY Susumu lai, Haj ime Tsuchida, and Setsuo Noda

ABSTRACT

One of the authors proposed a criterion in 1970

for assessing liquefaction potential. This criterion uses maximum acceleration of ground

surface and SPT N-value of subsoils around

ground water level. In the present study, the maximum acceleration is interpreted in terms of shear stress ratio and the SPT N-value of sub- soils in terms of so called corrected N-value.

Thus, the criterion proposed in 1970 is extended to incorporate the ability to predict

depth of liquefiable layers. The extended

version of the liquefaction criterion, namely

the New Criterion, is compared with the criterion proposed by Seed and Tokimatsu, et al. in

1984. They are in good agreement.

1. INTRODUCTION

Great earthquakes have caused damages to some

of the port structures in Japan. Among the various seismic effects to the port structures, liquefaction has caused one of the most serious

damages. Therefore, a due consideration should

be given for assessing liquefaction potential.

The Japanese port structures, the current

design standard specifies the following methods

for assessing liquefaction potential (1).

(1) a method using grain size accumulation

curve and SPT N-value,

(2) a method using cyclic shear tests of sand

specimen,

(3) a method using shaking table tests of sand

layers

.

Among these methods, the most suitable is chosen according to the degree of importance

given to the structures. For most standard

structure the method chosen is the grain size accumulation curve and SPT N-value. This method is based on a criterion for assessing

liquefaction potential proposed by one of the authors in 1970 (2,3). This criterion will be designated as "the 1970 Criterion" hereafter.

Several great earthquakes have occurred in Japan and caused liquefactions since the proposal in 1970. The liquefaction case records by these great earthquakes have not been contradictory to the 1790 Criterion.

However, the 1970 Criterion lacks the ability

to predict the depth of liquefiable layers because it uses N-values of subsoils located around the ground water level. It has been

sometimes difficult to decide how deeply the -liquefiable site should be improved. Therefore, the 1970 Criterion is extended to incorporate the ability to predict the depth of liquefiable layers. The report presents a method to extend the 1790 Criterion. The report also presents a comparison between the extended version of the 1970 Criterion, namely

the New Criterion, and the criterion proposed

by Seed and Tokimatsu, et al. in 1984 (15,17).

2. THE 1970 CRITERION

The 1970 Criterion is based on results of the shaking table tests and the liquefaction case records (2). Details were presented at the third joint meeting of U.S. -Japan Panel on Wind

and Seismic Effects, UJNR in 1971 (3). The procedure for assessing liquefaction

potential by the 1970 Criterion is summarized

as follows:

(1) Classify the soils by Fig. 1. Range (A): Easily Liquefiable Soil. Range (B) ((Bf) or (Be)]: Liquefiable

Soil. Other Ranges: Non-Liquefiable Soil.

(2) For soils in the ranges (A) or (B) , obtain

the critical N-value corresponding to the

maximxim acceleration of the design earth- quake by Fig. 2. For example, critical N- value is 16, if the soil is in the range

(A) and the maximum acceleration of the design earthquake is 300 gals. If the maximum acceleration is greater than 300 Gals, critical N-value is also 16 as indicated by the broken line for the soils

in the range (A) for most of the cases.

(3) Investigate the SPT N-value of the layer around the water level. Compare this N- value with the critical N-value. If the N-value is greater than the critical N- value, the ground is judged not to liquefy

at the design earthquake. If the N-value

is smaller than the critical N-value, the ground is judged to liquefy at the design

earthquake

.

Earthquake Resistance Structures Laboratory,

Structures Division, Port and Harbour Research

Institute, Ministry of Transport.

248
3. THE NEW CRITERION

3 . 1 Basic Rules of Liquefaction

The following basic rules of liquefaction have

become widely accepted among geotechnical

engineers in recent years.

(1) Liquefaction resistance depends on

looseness or denseness of soils.

(2) Shear stress ration r/a^^ (where r: shear

stress amplitude and ^vo- vertical effective pressure) is a good index of lique- faction resistance because value of r/a^^ which causes liquefaction does not depend

on

(3) The r/a^Q is greater, number of cyclic

loadings necessary for liquefaction

smaller. These basic rules are used for the extension of

the 1970 Criterion.

3 . 2 Relation Between SPT N-value and Effective

Vertical Pressure

As mentioned earlier, the 1970 Criterion uses

N-value' of the layer around ground water level. Effective vertical pressure of such layers in Japanese port area is about 0.66 kgf/cm^ . It

is known that N-value depends not only on

densities but also on effective vertical

pressure. Therefore, if the 1970 criterion has

to be applied to the layers at other effective

vertical pressures, N-values of those layers

have to be corrected for the difference in the effective vertical pressure. N-value corrected

into the value at the effective vertical pressure of 0.66 kgf/cm^ will be called "equivalent

N-value" hereafter and sometimes designated as (N)o.66- A relation between N-value and effective

vertical pressure for a fixed relative density

was studied for the sands shown in Fig. 3 in

the laboratory by several researchers (1,4-7). Ranges of the particle size distribution curves

shown in Fig. 1 are also indicated in Fig. 3. The results of the relation for sands at range

(A) are selected for the present study and are shown in Fig. 4 by broken lines. Average

relation is given by the following equation.

In addition to the laboratory study, there are several field investigations. Data shown in

Fig. 5(a) are the results by one of them. They

show a clear evidence of dependence of N-value

on effective vertical pressure. If the N- values before the excavation are corrected by

the equation (1), results become as shown in

Fig. 5(b). Therefore, applicability of the equation (1) is confirmed.

3 .

3

Relation Between Maximum Acceleration and

Shear Stress Ratio

If the 1970 Criterion has to be applied to the layers other than the layer around ground water

level, intensities of ground shaking can be better specified in terms of shear stress ratio

than maximum acceleration of ground surface as understood by one of the basic rules of lique- faction described in section 3.1. Relations between maximum acceleration of the ground

surface and maximum shear stress ratio within

the ground are shown in Fig. 6 by ground

response analysis of typical ground conditions

in Japanese port area as well as an empirical

equation (3). These results indicate that the layer around ground water level in the Japanese

port area receives a maximum shear stress ratio

of about 1.4 times larger as a^ax/S> where a^^x

is the maximum acceleration and g

is the acce- leration of gravity. Therefore, horizontal

axis of Fig. 2 can be interpreted as maximum

shear stress ratio if the value of maximum

acceleration is divided by 0.7 g. Instead of changing the horizontal axis of Fig.

2 into maximum shear stress ratio, maximum

shear stress ratio can be interpreted in terms

of equivalent acceleration if the maximum shear

stress ratio is multiplied by 0.7 g. Namely,

*ef 0.7 ( 2 )

where Ugf- equivalent acceleration in gals

’’max’ neximum shear stress in kgf/cm^ effective vertical stress in kgf/cm^

g; acceleration of gravity (980 gals)

In this way, the 1970 Criterion shown in Fig. 6 can be applied to any layers other than the layer around ground water level.

(N)o.66 N-1.828 (ct^^- 0.66) 0.399(a^'-0.66)+l

( 1 )

where (N)q self: equivalent N-value

N-value effective vertical pressure in kgf/cm^

Solid lines in Fig. 4 correspond to this

equation.

Concerning the influence of the number of cyclic loadings mentioned at section 3.1, there

was a proposal to determine the number of cyclic loadings by magnitude of the earthquake

(9). However, it is known that ground motions

are also greatly affected by other factors such

as response characteristics of subsurface

ground, distance from the epicenter, and source mechanism of the earthquake. Therefore, it is best to leave the influence of the number of cyclic loadings as an unknown factor and allow

249
a safety factor against this unknown factor for

the time being.

3 . 4 Additional Factors

From the discussion described so far, it is evident that the 1970 Criterion can be applied

to any layers other than the layer around

ground water level if N-value and maximum

acceleration of ground surface are interpreted

as equivalent N-value and equivalent acceleration. The knowledge acquired through recent

years has made it possible to discuss the following additional factors.

(1) Range below 100 Gals

The knowledge obtained so far for the liquefaction at very small intensity of shaking

is as follows:

a) There has been no case records of liquefaction below 100 gals in Japanese

port areas.

b) The shaking table tests show that several

sands liquefied at about 60 gals at their most loosely deposited conditions (2,3).

c) There is a report that sand will not liquefy if level of shear strain is below

10

- 2 % ( 10 ). If all these facts are considered, there may be

a limit of equivalent acceleration below which

no liquefaction takes place. However, at the moment, it is difficult to determine the exact

level. Therefore, for the time being, it is safe to draw a line from the origin of Fig. 2 to the end of the lines (A) and (B) [ (Bf ) or

(Be)] in Fig. 2 as a critical N-value line.

(2) Range over 300 Gals Recent studies confirm that dense sands do not liquefy and show a behavior so called cyclic

mobility. Therefore, critical N-value shown by

broken line in Fig. 2 can be reliable as a cri- tical N-value in this range. It may be better

to study the deformation of dense sands together with deformation of structures which

will exhibit a dynamic interaction with the sands by earthquake motions.

(3) Influence of Fines Content

The 1970 Criterion roughly expresses difference

of liquefaction resistances by particle size distribution curves shown in Fig. 1. Recent

studies show silty sands within range (A) in

Fig. 1 have lower critical N-value than clean

sands of the same liquefaction resistance.

Studies or specifications which show explicit

relation of fines content (less than 74 fi) are shown in Figs. 7 and 8 (11-14).

Whether the critical N-value line for the silty

sands can be obtained by parallel shift of the line or by magnifying the scale of N-value or acceleration coordinate remain an unresolved

question. For the present study, the authors propose using a reduction factor of N-value as shown in Fig. 9, which is roughly harmonious to the results shown in Figs. 7 and 8.

(4)

Allowance for the Possible Uncertainty of N-values and Others

From the discussions presented so far, a basis

for the New Criterion has been established. Liquefaction potential will be determined by

the comparison of equivalent N-value to the critical N-value which corresponds to the equivalent acceleration. However, if N-value

is very close to the critical N-value, further

studies on the liquefaction potential will be necessary because N-values always contain

uncertainty as a representative indices of the liquefaction resistance at the ground layer under investigation. Moreover, influence by

the number of cyclic loadings by the earthquake

motions are not explicitly taken into account.

The regions of the uncertainties can be expressed in terms of safety factor for equivalent acceleration F2 for the range of equivalent acceleration less than about 300 Gals and equivalent N-value less than about 16. It is difficult to specify the exact values of

F2 for the region of uncertainties but, roughly

speaking, the region of uncertainties is considered 1/1 . 5<F2<1 . 5 . Fig. 10 shows this

region.

For the range of greater equivalent acceleration and greater equivalent N-value,

the safety factor for equivalent F2 is not

longer effective as understood by Fig. 10 because horizontal broken line of (N)q gg-16 is the critical N-value line in this range and it

is parallel to the equivalent acceleration

axis. Therefore, it is necessary to allow a safety factor for equivalent N-value in this

range. The study of Tokimatsu and Yoshimi

shows that, if N^ (N-value at effective

vertical pressure of 1 kgf/cm^ ) is greater than

25, it is safe from liquefaction (12). This value of Nj^ corresponds to 20 in equivalent N- value in this study. Therefore, (N)g gg-20 is taken as a border line of the uncertainty

region in this study.

3 . 5 The New Criterion

The New Criterion is summarized as follows:

(1) Soils are classified by Fig. 1.

(2) Equivalent N-value is calculated by

equation (1) for soils in the ranges (A) or (Be) in Fig. 1. For soils in the range

250
of (Bf) , no correction will be applied

and N- value is considered to be the equivalent N-value.

(3) Equivalent acceleration is calculated by

equation (2) , where maximum shear stress

ratio is computed by ground response

analysis

.

(4) Liquefaction potentials of soils in the range of (A) in Fig. 1 are judged by Fig. 11(a) and those in the range of (Bf) or

(Be) are judged by Fig. 11(b). If the layer of the soils in the range of (Be) is overlain by impermeable layers, its liquefaction potential will be judged by

Fig. 11(a) instead of Fig. 11(b).

(5) For the soils in the range of (A) in Fig.

1, critical N-value and associated boun- daries of uncertainty regions are reduced

by the reduction factor shown in Fig. 13 if fines content is greater than 5%.

the Criterion by Seed and Tokimatsu, et al .

,

hereafter. This criterion was also presented

at the 17th Joint Meeting of U.S. -Japan Panel

on Wind and Seismic Effect, UJNR and in the reference 17 in 1985 (16). The New Criterion

proposed in the present paper is compared with

the Criterion by Seed and Tokimatsu, et al., in this chapter.

4 . 1 Dependence of N-value on the Effective Vertical Pressure

Dependence of N-value on the effective vertical

pressure is shown in Fig. 12, in the normalized

form by the equivalent N-value. Parameters used in the Criterion by Seed and Tokimatsu, et

al., are relative densities, but those used in the New Criterion are equivalent N-values. Therefore, precise comparison is not possible.

However, roughly speaking, they are in agreement

.

4.2 Critical N-value

(6)

From the Fig. 15, following judgments are obtained. Region I: Liquefaction takes place. Liquefaction is judged to take place

.

Region II: Liquefaction likely takes place

therefore, liquefaction is judged

to take place, or further study such as laboratory tests of soils

is recommended if improvement of liquefiable soils is economically

or technically difficult. Region III: Liquefaction does not likely take place. Therefore, liquefaction is judged not to take place; labora- tory tests of soils is recommended

if the structure is considered a critical facility.

Critical N-values used by the criterions are shown in Fig. 13. The horizontal axis of Fig.

13 shows N-value at effective vertical pressure

of 1 kgf/cm^ obtained by the SPT procedure with

rod energy ratio of 60% using a constant 1-3/8" diameter barrel. Because the rod energy ratio

of Japanese SPT procedure is 78% by mechanical

trip system (Tonbi) and 67% by rope and pulley

system (15), average rod energy of 73% is considered to be the basis of the New Criterion

in which no distinction is made of the methods

of SPT. The New Criterion is shown in Fig. 13 taking these facts into account.

In overall, the criteria agree well with each

other

.

The following conclusions are obtained from the

s tudy

.

Region IV: Liquefaction does not take place. Liquefaction is judged not to take place

.

Extensive study on the applicability of the New

Criterion has been carried out by using lique- faction case records in Japan. This part of

the study will be reported in a separate

report. The result of the study has confirmed

the applicability of the New Criterion.

4. COMPARISON WITH THE CRITERION BY SEED. TOKIMATSU. ET AL. (1984)

Seed, Tokimatsu, Harder, and Chung have proposed a Criterion in October 1984, based on

the comparative study of standard penetration

test procedure among U.S., Japan, Argentina,

and China (15). This Criterion will be called

(1) A criterion proposed by one of the authors

in 1970 can be applied to the layer located around ground water level but not

to the other layers. The present study

makes it possible to apply the criterion

to all the layers and thus makes it pos- sible to predict the depth of liquefiable

layers

.

(2) The criterion thus obtained was compared

with the criterion proposed by Seed and

Tokimatsu, et al. in 1984. The results of

the comparative study of SPT procedure between U.S. and Japan were taken into account for the comparison. They agree

well with each other.

251
6. REFERENCES

1) Bureau of Ports and Harbours, Ministry of Transport: Design Standard for Japanese

Port Facilities and the Guidance, Japan

Port Association, March 1979 (In Japanese)

.

2) Tsuchida, H. , Prediction of Liquefaction

of Sandy Subsoils and Countermeasures to Possible Damage Caused by the Liquefaction, Proceedings of the Annual Meeting of

the Port and Harbour Research Institute, December 1970 (in Japanese).

3) Tsuchida, H. and S. Hayashi, Estimation of Liquefaction Potential of Sandy Soils, Third Joint Meeting of U.S. Japan Panel on Wind and Seismic Effects, UJNR, May 1971.

4) Gibbs, H. J. and W. G. Holtz, Research on Determining the Density of Sands by Spoon Penetration Testing, Proceedings of the Fourth International Conference on Soil Mechanics and Foundation Engineering,

London, 1957, Vol. 1, pp. 35-39.

5) Marcuson, W. F. Ill and W. A. Bieganousky,

Laboratory Standard Penetration Tests on

Fine Sands, Journal of Geotechnical Engi- neering Division, ASCE, Vol. 103, No. 6, June 1977, pp. 565-588.

6) Marcuson, W. F. Ill and W. A. Bieganousky,

SPT and Relative Density in Coarse Sands, Journal of Geotechnical Engineering Division, ASCE, Vol. 103, No. 11, November

1977, pp. 1295-1309.

7) Yoshida, Y. , G. Kokusho, Y. Esashi, M. Sasaki, Standard Penetration Tests on Dense Sands (Second Report) , 18th Japan

Geotechnical Engineering Symposium, June

1983, pp. 45-46 (in Japanese).

8) Tatsuoka, F. , T. Iwasaki , K. Tokida, S. Yasuda, M. Hirose, T. Imai, and M. Konno, Standard Penetration Tests and Soil Lique- faction Potential Evaluation, Soils and Foundations, Vol. 20, No. 4, December

1980, p. 95-112.

9) Seed, H. B. and I. M. Idriss, Simplified

Procedure for Evaluating Soil Liquefaction

Potential, Journal of the Soil Mechanics

and Foundation Division, ASCE, Vol. 97, No. SM 9, September 1971, pp. 1249-1273.

10) Dobry, R. , R. S. Ladd, F. Y. Yokel, R. M. Chung, and D. Powell, Prediction of Pore Water Pressure Buildup and Liquefaction of Sands During Earthquakes by the Cyclic Strain Method, NBS Building Science Series

138, U.S. Department of Commerce, National Bureau of Standards, July 1982, 50p

.

11) Ishihara, K. , Stability of Natural

Deposits During Earthquakes, Proceedings

of 11th International Conference on Soil Mechanics and Foundation Engineering, August 1985, pp. 321-376.

12) Tokimatsu, K. and Y. Yoshimi, Criteria of Soil Liquefaction with SPT and Fines Content, Proceedings of Eighth World

Conference on Earthquake Engineering, July

1984, Vol. 3, pp. 255-262.

13) Fire Defense Agency, Regulations on Controlling Hazardous Materials, 1978 (in Japanese)

.

14) Japan Gas Association and Committee on Safety of LNG Tanks: Provisions on LNG Underground Tanks, 1979 (in Japanese).

15) Seed, H. B. , K. Tokimatsu, L. F. Harder, and R. M. Chung, The Influence of SPT Procedures in Soil Liquefaction Resistance

Evaluations, Earthquake Engineering Research Center, University of California,

Berkeley, Report No. UCB/EERC - 84/15

,

October 1984, 50p.

16) Chung, R. M., H. B. Seed, A. G. Franklin, and J. A. Farrar, Re-evaluation of Liquefaction Resistance Curves, the 17th Joint Meeting of U.S. -Japan Panel of Wind

and Seismic Effects, UJNR, May 1985.

17) Seed, H. B. , K. Tokomatsu, L. F. Harder, and R. M. Chung, Influence of SPT Proce- dures in Soil Liquefaction Resistance

Evaluations, Journal of the Geotechnical

Engineering, ASCE, Vol. Ill, No. 12, December 1985, pp. 1425-1445.

252
(a) For Soils of Uniform Granding

(b) For Well Graded Soils Fig.l Ranges of Particle Size Distribution Curves

of Liquefiable Soils

Max . Accel. ( Gal

)

Fig. 2 Critical N-values for Assessing

Liquefaction Potential

253
Percenloge

Finer

by

Weight

Percentage

Finer

by

Weight

Particle Size 0 (mm) Fig. 3 Particle Size Distribution

Curves of Sands Used for Standard Penetration Tests

In the Laboratory

Fig. 4 Relation Between SPT N-value

and Effective Vertical Stress from the Laboratory Test Data

Ground N-value

(a) N-values Before and After

the Excavation

(b) Computed N-values and Those Measured After the Excavation

Fig. 5 Comparison of N-values Before and After the Excavation
Reduction

Factor

for

Critical

N-volue

Normalized Max. Shear Stress Ratio (^)/(^)

0.5 1.0 1.5

10

20

30

.ii 40

50-

60

<j; g u;

Z: Depth From the Ground Surface

(Level of Ground Water ; 2m)

•(I-Q0I5Z)

10 / • o/

\,Q-

*

[ ,10 r

Fig. 7 Increment of N-value in Accordance with Fines Content

® 25

> 20

^ 1510 - 5

_ Japanese Regulation for Oil Tanks

‘(Vicinity of Oil Tanks) Japanese Regulation for Oil Tanks (Under Oil Tanks)

. Japanese Provision forLNG Tanks

0 5 10 15 20 25 30 35

Fines Content (%)

Fig. 6 Distribution of Maximum

Shear Stress Ratio Normalised

by Maximum Acceleration of Ground

Fig. 8 Provisions Explicitly Taking

Into Account the Influence of Fines Content in Terms of Variation in Critical N-value

Fig. 9 Proposed Reduction Factor for Critical N-value

Fig. 10 Critical N-value Line and Associated Uncertainty Regions

255
(a) For Soils in Range (A) in Fig.l

Fig. 11 Proposed Chart for

Equivalent Acceleration Oeq (Gal) For Soils in Range (Bf) or (Be) in Fig.l

Assessing Liquefaction Potential

(b)

N/(N)o66

I—^Seed , Tokimotsu.et al.

( 1984 )

Inter- No Significont

U'UIIIUUC - Potent, nl mediate Damage

Fig. 12 Comparison with the Criterion

Proposed by Seed, Tokimatsu,

et al. for the Relation of N-value and Effective Vertical

Fig. 13 Comparison with the Criterion

by Seed, Tokimatsu, et al

.

for Critical N-value (Clean Sands

)

Pressure

256
EARTHQUAKE SAFETY INVESTIGATION OF

EXISTING FACILITIES

BY Joseph V. Tyrrell, P.E. ^

ABSTRACT 2. PROGRAM irCPLEMEHTATION The U. S. Navy has a large number of bases

located in areas where seismic risk is high. Many bases are in costal areas and areas founded on weak soils and hydraulic

fills. Experience from recent earthquakes

has led to increased design requirements and

some concern for the safety of existing

structures. Pilot studies have shown that

about 5% of the structures at these bases

are liable to experience serious damage. To deal with the large inventory, at a reasonable cost, a systematic investigation

procedure was developed . This paper

describes the methodology adopted.

KEYWORDS: earthquaike safety; existing

facilities; earthquake investigations;

seismic

1. INTRODUCTION

The U.S. Navy has 55 major bases, as well as a scattering of small facilities,

located in areas of high seismic risk in the

50 states, Puerto Rico, Guam, the Philippines, and Naples, Italy (included because of special problems). Other

overseas areas are not being investigated at this time. This includes utilities, roads, bridges, dams, and over 14,000 major

buildings. Even a superficial investigation

of existing structures on this scale could

be costly. Therefore, a strategy to reduce

the scope of the investigation was needed.

Several possible approaches were

considered. One plan was to limit the investigation to the essential facilities.

This was discarded because we were concerned

with personnel safety as well as the facility function. A second candidate

approach was to have experts scan the facilities on-site and select buildings or other facilities which require

investigation.

A field screening by experts was incorporated as one step in the preliminary

work, but after the scope was first reduced

by computer screening. The scheme adopted

utilizes several layers of screening and two levels of evaluation in order to minimize

the effort and its cost.

The investigation program is handled by the Naval Facilities Engineering Command's

(NAVFAC) six Engineering Field Divisions,

each of which serves a specific geographical

area. About 90^ of the work is done thru contracts with private A&E firms. NAVFAC HQ provides funding and policy guidance. The Western Division, located in San Bruno, CA. assists NAVFAC HQ with coordination and planning. The Naval Civil Engineering

Laboratory (NCEL), Port Hueneme, Ca. has provided support in the form of research, development and consultation. The final product is recommendations for needed improvements and draft programming

documents. The operating Commands at the various naval activities determine what type

of funding is applicable and integrate

earthquake improvements with their other requirements

.

3. METHODOLOGY

The various steps in the evaluation process are discussed in this section. Often it is convenient to combine several of these steps into one operation. A new tri-service design manual for investigation of existing facilities is now being prepared. The Navy's designation for this manual will be NAVFAC P 355.2 (ref. 1). The procedures given in this manual are essentially the same as those being used in the Navy's Earthquake Safety program which consists of the following steps:

a. Inventory Reduction (computer screening)

b. Consultation with the Activity

c. Site Seismicity Studies

d. Evaluation of Site Hazards and Utilities

e. Field Screening and Data Collection for Structures

f. Preliminary Structural Evaluation

g. Preliminary evaluation report

h. Follow-up Investigation of Site Hazards

& Utilities

i. Detailed Analysis of Structures

(identified by Preliminary Evaluation)

j. Recommendations for Mitigations (if Requi red)

^ Naval Facilities Engineering Command, Consultant in Earthquake Engineering

257
k. Final Evaluation Report (with cost and programming data). Steps one and two are usually completed

before a contract for an investigation is prepared. In most areas, the balance of the work is divided into two phases which are contracted for sequentially. Results from

the first phase define the scope of work

required in the second phase. The process

continues to evolve and consequently not all studies have been done in exactly the same manner.

3.1 Inventory Reduction

The Navy is fortunate to have a good deal of information on its facilities in an

automated data base. The real property

listing provides useful information on the

age, size, cost, and tjrpe of construction.

As a first cut, buildings with the following

attributes are eliminated:

a. Constructed after 1973

b. Area less than 3000 SF

c. Single story pre-engineered metal or

timber structures

d. Structures with replacement cost less than $200,000

e. Earth covered structures, (6) structures

scheduled for replacement within 5 years

f. Family housing.

This screening usually eliminates more than

half of the structures. In addition, where

there are identical structures, we try to limit the investigation to just one since

the results may be inferred for the others.

A few of the structures eliminated may be

later reinstated as discussed below.

3 .2 Consultation with Activity

The initial list produced by the computer

screening is presented to the Public Works group at the Naval Activity in question.

They can effectively identify errors in the computer listing and buildings that may not warrant investigation for various reasons. Sometimes the Activity will request investigation of structures not included after the initial screening. We either add these candidates or explain why it is unnecessary.

3 .3 Site Seismicity Studies

A seismicity study is made to establish site specific response spectra. This is used is conjunction with the rapid evaluation method described below. It is also used in conjunction with the design of new essential

facilities in accordance with the new tri-service design manual. Seismic Design Guidelines for Essential Buildings [_ref 2]

.

3 .5 Evaluation of Site Hazards 8c Utilities

This is usually done in conjunction with a site visit. Visual inspection and discussion with the public works engineers

are the best sources. When appropriate,

soil reports for existing construction and utility maps are scanned. At this point in the investigation, attention is directed

toward the overall utility systems, not utilities inside of buildings. The reliability of utilities supplied from outside sources is difficult to evaluate but

is essential in deciding what on-base capacity is needed for emergencies,

3 .6 Field Screening & Data Collection

NAVFAC or A&E engineers making the evaluation visit the site and visually

screen the structures. Some structures usually can be eliminated and sometimes an additional structure may be recommended. In some cases the need and type of upgrading

required may be obvious, and this may be noted without further effort. When a listed structure is found to be a valid candidate,

a walk thru inspection is made and notes taken with regard to hazardous non-structural components as well as pertinent observations on the structural

system. The condition of materials should be noted. Available drawings and calculations are reviewed or collected for subsequent review in the office. Photos are useful for later reference.

3 .7 Preliminary Structural Evaluation

In this phase any necessary data not previously collected is obtained and the information is evaluated. The "rapid analysis" procedure developed by the firm of John Blume/URS (ref 3) is applied to assess vulnerability. In this method the key structural elements are identified and their capacities to resist earthquake forces estimated, a typical summary sheet is shown

in Table 1 at the back of this manuscript. A resistance curve for the structure can then

be compared to the earthquake demand.

Demand is determined from site response spectra (for various degrees of damping)

developed by a seismicity study. The calculations can be done fairly simply

and the results quickly obtained by a graph

showing the demand and resistance lines, see Figure 1 at the back of the manuscript. The ordinate is the spectral acceleration and the abcissa is the period. When the initial yield is reached, damage is assumed to be zero. When the ultimate capacity is reached (collapse mechanism developed) damage is assumed to be 100^. The period of the structure will lengthen and its ductility

changes as it experiences inelastic deformation.

258
This evaluation method assumes that damage

varies linearly between yield an ultimate.

A point is found on the line connecting Py. and P^ such that the horizontal distance

from Py is to the horizontal distance to P^ inversely as the horizontal distances

from the point to intersections with the demand spectra for elastic damping and damping at ultimate. "Percent of damage" is defined as the horizontal distance from

yield to the point divided by the distance

from yield to ultimate.

The Naval Civil Engineering Laboratory

(NCEL) has provided modifications of the Blume/URS methods and developed computer

programs for its application. This paper is not intended to provide a complete

description of the rapid analysis

technique. Anyone interested in the details

of this method should obtain a copy of

references 1 & 2. The method is for preliminary evaluation

only. It identifies which buildings are

liable to be damaged and gives an indication

of their relative vulnerability. It cannot

be used for all structures.

3

.8

Preliminary Evaluation Report

At this point in the procedure, it is necessary to determine what additional

investigation is needed and its cost. Possible items include follow-up on site hazards and utilities, and detailed

investigation of buildings which appear to

be very vulnerable. As a rule of thumb, we consider buildings for further investigation

if the rapid analysis indicates damage greater than 60%. Essential buildings with

indicated damage greater than "30% may

warrant further investigation.

Buildings that are subject to potential

damage from site hazards may not be worth

further structural analysis unless mitigation of the site hazard is feasible. Buildings with severe foundation problems

may not be appropriate for further study.

This report also includes findings with

respect to non-structural elements and recommendations for mitigations of non-structural problems with order of magnitude costs. It does not contain an itemized listing of every component although

some major pieces of equipment may be described. In general, broad categories

such as light fixtures are discussed. The investigator should provide sketches showing remediaL measures.

3.9 Follow-up Investigation of Site Hazards

& Utilities

The investigation of site hazards depends upon the unique conditions prevailing at each site. It might be necessary to determine the boundaries of area subject to tsunami, liquefaction, or slides. Gas lines

and storage of hazardous materials pose a threat of secondary damage. Looping and redundancy in utilities provides increased confidence that operations will not be disrupted and that recovery can be carried

out expeditiously.

3 . 10 Detailed Analysis of Structures Structures recommended for further investigation after the preliminary

evaluation are reviewed and a final list approved. The investigator may use whatever

methods he feels necessary and appropriate.

At this point it is not necessary to analyze

the entire structure and determine stresses

in each member. Instead an analysis that deals with typical segments of the lateral

force resisting system is appropriate. It is our aim to do as much or as little analysis as is required to determine if deficiencies exist and their nature and extent. The need for upgrading is determined

.

3 .11 Recommendations for Mitigation

Under this task, the detailed analysis is continued to determine what effective mitigations are possible and their cost. Usually the results for typical lateral

force resisting elements can be used to establish a budget estimate for upgrading.

The first mitigation considered should allow

the structure to meet 100^ of the demand, unless this is not cost-effective. The minimum acceptable level of safety is that the structure's capacity be equivalent to at least 73 % of the earthquake demand. If this cannot be accomplished, the report will usually indicate that upgrading is not feasible. However, if a leaser tangible

improvement can be made at small expense, we will consider it. We rely on the investigator to determine whether there is more than one alternative worth considering.

3 .12 Final Evaluation Report

The final report is expected to summarize

all the investigations and present the recommended mitigations and their cost for

the selected buildings. It also includes

drafts, in the correct format, of documents

that the Activity can use to request funding

for remedial projects.

259
4. ECONOMIC ANALYSES

In order to justify upgrading, some form economic rationale or cost/benefit analysis

is often required. ¥e can’ evaluate structural effects fairly accurately

although there is some uncertainty. Other

elements of cost are more difficult to deal

with. These include damage to ncn-structural elements, damage to equipment

and building contents, secondary damage

(fire, etc.), as well as injury of personnel. Additionally there is the loss

of productive capacity while repairs are

made. For the Navy, there is the question

of military readiness. These factors are

not amendable to precise quantification,

however, an economic analysis is incomplete

and may be misleading if they are not included

.

5. PRIORITIES FOR UPGRADING When there are a large niimber of facilities

requiring upgrading, it is necessary to establish priorities. Some of the factors

to be considered are:

a. Vulnerability of the structure

b. Vulnerability of contents

c. Feasibility - Cost /benefit of upgrading

do Importance of facility - impact of loss

So Occupancy (no. of people)

fo Expected service life of the facility

Requirements for upgrading for earthquake

safety must, of course, be integrated with

other requirements. The Navy spent some time considering systematic methods for prioritizing projects, but without complete

success.. However, each activity provides

the prioritization for projects it recommends for funding, and at this level,

no activity has so many upgrading projects

that it requires a structured system. On further reflection, it is also apparent that an individual facility might be upgraded but still not be able to function because of damage to associated facilities. In order

to substantially affect military readiness,

a broad improvement program is needed.

The following priority order is suggested as most effective for improving overall safety

as soon as possible:

a. Facilities needed for damage control and recovery

b. Utilities

c. Structures liable to collapse

d. Non-st ructural hazards (low cost items such as light fixtures, etc.)

e. Structures subject to serious damage

f. Essential structures subject to moderate

damage

g. All other

6. STATUS & RESULTS Investigations thru preliminary evaluation

have been completed for about 90^ of the facilities. The final reports have been

produced for about If funding permits

this pace to be continued, the investigations should be substantially

finished by 1989. A total of 55 bases and more than 14,000 buildings will have been

considered. The total cost is estimated to be 14 million dollars of which approximately

10 million is attributable to evaluation of structures. Site hazards, particularly

liquefaction, have been identified at about

10 bases. About 3 % of the existing

structures, approximately 750, require structural upgrading. Naval Activities have implemented expedient

safety measures at many bases. These

efforts have focused on minor utility

improvements and non-st ructural elements. NAVFAC has issued a directive with regard to modernization and repair projects which has resulted in seismic upgrading of at least

130 buildings (some not yet fully completed). The general investigation of existing structures has resulted in upgrading of about 60 additional buildings with about 200 projects pending. At this point it is not possible to say what percentage of the deficient structures are likely to be upgraded.

The upgrading required may be anything from

a few X braces to a completely new lateral

force resisting system. Consequently, we have not been able to establish any accurate

estimate of construction costs. A projection of an early sample indicated that total construction costs would be on the order of 200 to 3OO million (not including

hospitals) . A decision to undertake or dismiss a few major projects could strongly

affect this figure. The cost of investigation though preliminary evaluation has been about $1,700 per building listed in the evaluation report (much less for the total buildings

screened). The cost of detailed evaluation may vary widely depending upon the complexity of the building but has averaged

about $25,000 per building. Overall cost is about $1,000 per building screened, but this figure is misleading because it includes

site investigation of site hazards and utilities

.

The most ambitious seismic upgrading project undertaken up to this time involves the Naval Hospital located in Oakland, California. Design for a modification to this 500 bed hospital with clinics and support facilities is complete and

260
construction is pending. The portion of

cost associated with seismic upgrading is estimated to be about 27 million dollars. Another pending project involves the Naval

Support Activity at Naples, Italy. The site presently occupied is exposed to both

tectonic and volcanic activity. For the

past two and one half years, uplift

associated with a magma intrusion and accompanied by tremors has been causing

minor damage. The present facilities are

leased. To construct facilities at a better

site is estimated to require funding on the order of two hundred million dollars. Design

for some of these facilities is underway.

7. SUMMARY

A methodology for investigating earthquake

safety for a large inventory of facilities

has been developed and implemented. Cost is minimized but still substantial, however, it

is small compared to the potential losses which would result from a major earthquake.

The method is adaptable to a variety of requirements and purposes. A user can customize the method by changing the factors

used for screening and evaluation. Each

user is likely to have unique situations

which determine how facilities are prioritized for upgrading.

In the event that the accurate prediction of

earthquakes becomes a reality, the identification of facilities subject to serious damage would still be needed as a part of any warning system.

This type investigation has been feasible

for the Navy and we feel that it will be

cost effective. Applications in the public domain would involve many additional

problems as well as political questions with

respect to implementing mitigations. A

broad upgrading of existing structures

involves large expenditures and consequently

warrants a plan to achieve the most cost effective results. Judging by the Navy's experience, the extent

of upgrading needed in the USA and world-wide necessitates accomplishment over

an extended period of time. On the other

hand, if a future large earthquake should

cause massive damage, a crash program for upgrading may be the result

.

REFERENCES

1.

Seismic Design Guidelines for Upgrading

Existing Buildings, TM b-809-10-2/NAVTAC P 355.2/AFM 88-7, Chapt . 13.2 , Departments of

the Army, The Navy, and The Air Force, Washington, D.C.

2 . Seismic Design Guidelines for Essential

Buildings, TM 5-809- 10-2/NAYFAC P 35b .T/aFM 88-3 » Chapt. 13.1 , Departments of Army, The

Navy, and The Air Force, Washington, D.C.

3 . Puget Sound Naval Shipyard ; Dynamic

Analysis, Rapid Evaluation Method, and Coat Estimates for Damage, by Sigmund A. Freeman, Proceeding of the NAVFAC Structural Engineering Workshop, 1974.

261
TYPE A STEEL STRUCTURES , BUILDING INITIAL YIELD ULTIMATE

SEQ. NAVFAC

S’a S'd NOTE

NO. NO. DIRECT. TYPE T ^'d ELEMENT T ^‘a ELEMENT REF.

17 462 L A2 0.25 0.40 0.2 -X 1.2 0.54 7. C & X T 0.9 0.50 4. c 1.1 0.57 6. CSX

18 469 L A2 0.25 0.17 0.1 X 0,70 0.42 2. C & X T A4 0.45 0.40 0.85 c

19 494 L A2 0.25 0.40 0.2 X 1.2 0.54 7. C & X 3 T 0.9 0.50 4. c 1.1 0.57 6. C & X 20 818 EW A2 0.2 1.2 0.2 X 1.1 0.45 5. c NS 0.1 2.4 0.1 X 0.45 1.6 3. c

21 851 L A2 0.25 0.29 0.2 X

. = - T 0.35 0.14 0.2 X 1.5 0.27 5. c

TABLE 1

T

S' Spectral aoceleration capacity in g's S'j Spectral dlsplaoanesit capacity (relative to ground) in inches

Eleraent Syntolas C CDliimj ® Nonstructuralj X Diagonal Bracing B BriOt wall: W Hood wall; T Tile wall Building Types: PC Mill bldg, with diagonal bracing

A3 Hill bldg, with diagonal bracing

A4 Mill bldg, without bracing or walls

Mapped fma reference J

Assumptions: Damage Ot 9 Damage 1001 0

X Damage Linear Between %

‘

’’u Damping, Linear Between

P^ (2t) i Py (5X)

Example: Earthquake Criteria for

101 Probability of Being Exceeded, Damping at

21 and 5% (Appendix C)

Steel Building {«469L)

2% Damping 0 Yield Capacity. P^

51 Damping 0 ULT. Capacity, P^ (Table 0-1. Appendix D)

Period (T), Sec.

FI(?URE 1 Adapted from ref, 3

262
THEMATIC MAPPING WORKS FOR SEISMIC DISASTER PREVENTION

IN GEOGRAPHICAL SURVEY INSTITUTE, JAPAN

BY Masatoshi Nagaoka and Hiroyuki Matsuda L

ABSTRACT

The Geographical Survey Institute investigates earthquake disasters and landform conditions

after major earthquakes occur. Based on the results of these investigations, the facts and findings about the relation between earthquake

disasters and landform conditions of lowland

are explained in this paper. The knick line

was identified to have a relationship to slope

failures. In addition, the viewpoint of historical evolution (development) of landforms

is adopted to explain the increase of slope instability after investigating the Naganoken-Seibu Earthquake of 1984 (Ontake Landslide , etc . )

. The Geographical Survey Institute is engaged in various geographic works Including thematic map

making of major areas in Japan, such as the Land Condition Map that addresses landform classification and the Land Use Map that addresses land use and urban function. This work is expected to contribute to the prediction of earthquake damages.

KEYWORDS: Earthquake disasters, Landform

influence. Thematic mapping

1. INTRODUCTION

Earthquakes have distinguished characteristics

in frequency and Intensity. There have been many efforts to predict earthquakes scientifically and take measures against them administratively. The great Kanto Earthquake

of 1923 and Fukui Earthquake of 1948 are two representative examples of disastrous earthquakes that affected big cities. Today,

if large cities were subjected to such earthquakes, the damage could be unpredictably

severe due to the congestion of people and buildings and falling debris.

To implement effective measures against earthquakes, it is necessary to be able to predict the location of the disaster-stricken

area, the type of disasters, and their intensity. Many investigations have been performed on understanding the relationship between natural condition of land and earthquake disaster in Japan. The Geographical Survey Institute performed investigations on all major

earthquakes since 1948. Earthquake damage

varies sharply with landform and ground

condition. It is effective to know ground condition in the plain by landform type based

on landform classification, because classified landform type is closely related to the historical evolution (development) of landforms

and is related to ground condition.

2. INVESTIGATIONS ON EARTHQUAKE DISASTERS

PERFORMED BY THE GEOGRAPHICAL SURVEY

INSTITUTE

The Geographical Survey Institute has performed post-disaster investigations of damages caused

by major earthquakes since World War II. Included in these investigations are the relation between the damages and land conditions. Thematic maps were produced as Investigative results. The results of the survey are discussed below.

2 . 1 Fukui Earthquake of 1948 An organized survey on earthquake damages and landform conditions was performed after the Fukui Earthquake in 1948. The following are

the conclusions (reference 1): a. Earthquake damages centered in alluvial

lowland. Damage ratio of wooden houses was related to the thickness of soft ground.

b. Most of the destruction resulted from slippages of deep faults and accompanying earth movements

.

c. Location of earthquake faults and earth movements could be estimated to some degree if tectonic landforms in alluvial lowlands were researched.

2.2 Chilean Earthquake Tsunami of 1960 Results of Investigation on major tsunami-stricken places revealed that the scale

of damages by this tsunami varied with

regions. This regional difference in damages

was found to be related to the difference in coastal landform as shown below (references 2 and 3):

Geographical Survey Institute, Ministry of Construction, Kitasato, Yatabe-machi

,

Tsukuba, #305, Japan.

263
a. Even along the Rlas Coast, damage intensity

by the tsunami differed from place to place; it varies with submarine topography. The type of tsunami varied with the shape of the bay, such

as the V-shaped bay (a submergence valley of a small river of the past), the W-shaped valley

(a complex of V-shape bays), and the U-shaped

bay (a submergence bay of a big river in the past)

b. Routes of intrusion, maximum reach from the shore and length of inundated period are deeply

related with the landform of coastal plains.

c. By classifying shoreline and landform of coastal plains on the basis of the above, a regionally different type of tsunami and its risk can be estimated. In this sense, a landform survey on the coast forms give the basis of measures against a tsunami.

2.3 Niigata Earthquake of 1964 Unique damage from liquefaction at alluvial

soft ground occurred following this earthquake. The investigation of the damage and landform condition (aerial photographs, interpretation of landform, field survey, mapping of the damage, etc.) resulted in the following conclusions (references 4, 5 and 6):

a. Damages were serious along the former river

beds. Also, major damage was found at Interdune lowlands, backmarshes and banked-up

lands (former lowlands or backmarshes).

obtained (reference 7): a. Landslides frequently occurred around the knick line in mountain slopes.

b. In the relationship between landslide and dip division, it was revealed that the occurrence of landslides increased rapidly at 45 deg or greater slope according to landslide occurrence ratio by every dip unit. At slopes overlaid by tuff and/or pumice layer, however, landslides occurred even at about 25 deg. 2.5 Mlyagi-Ken-Okl (Off the Coast of Minagi-Pref) Earthquake of 1986 The damage centered on newly prepared housing

sites overlying hills. The following facts were revealed in connection with landform

(reference 8): a. Thicker banked-up layers resulted in greater damage to houses and ground. Sizable damage occurred at places where the slope before banking was large and where a terrace-like banked-up slope was one-sided.

2 .6 Nlhonkai-Chubu (Central Japan Sea) Earthquake of 1983 The following, similiar to preceding cases, were revealed (reference 9): a. The highest percent of damage occurred to wooden houses at the banked-up former marsh.

b. Landform on ground surface is classified by characteristics such as geomorphic agent, formative age, shape and relief, structure, composed material, etc. The landform unit classified revealed that it is not only

characteristic of landform itself but also the relation with surface geology and soil and drainage condition. Such detailed classification of landform, therefore, explains

the reason of occurrence of regional difference

in damage

b. Damage percentage was relatively high at the cut and leveled higher sand dunes, lower

sand dune (groundwater level is high), inter-sand dune lowlands, banked-up flood plains and former river beds.

c. Damage percentage at a terrace and a table

land was lower than that of any other type of landform.

2.7

Naganoken-Selbu Earthquake of 1984

c. Consequently, earthquake damage potential

at another place with similiar characteristics

can be predicted to some degree by a microlandform classification map.

d. From the above, it was proven that a Land Condition Map which included damage made by the Geographical Survey Institute, was effective in predicting earthquake damage potential.

2.4 Izu-Hanto-Okl (Off Izu-Penlnsula) Earthquake of 1974 and Izu-Oshlma- Klnkai (Adjoining Sea of Izu-Oshlma) Earthquake of 1978

The damage from this earthquake was reported at the 17th U.S. -Japan Joint Meeting (UJNR) in 1985. The author reported on geomorphological

and geological characteristics of a large-scale

landslide (Ontake Landslide), which occurred at the same moment when this earthquake occurred.

That is, 3.4 x 10^ m^ of mountain mass suddenly began landsliding. This is shown in Fig. 1 as landform of prelandslide and postlandslide (stereoscopic). The main mass landslided along underlying pumice layer, and resulted in the reappearance of a V-shaped buried valley (reference 10). Topographic profiles of landform change and landform classification maps of this landslide were produced. From the viewpoint of Quaternary historical evolution (development)

After the investigation of the relationship

between frequent landslides and local landform

condition, the following conclusions were

264
of landforms, the author studied:

a. Why did this large-scale landslide occur only at that site although the earthquake and heavy rainfall existed in the neighborhood?

b. Why, for the first time, did the landslide

occur at the site despite the fact that no earthquake had been accompanied by a landslide

before?

By reviewing historical factors connected to the occurrence of landslides, the following

finding is obtained: a landslide occurs when

the Instability Index I (reciprocal of "factor

of safety") (reference 11) exceeds the critical

value. The Instability Index is shown below.

I

. i X CVj X A + + V^) where:

I = Instability Index (reciprocal of "factor

of safety").

L = Land relief factor. M = Slope material strength factor.

V, = Final forcing variable for generating of landslide. A = Amplification coefficient for by

site condition.

Vg = Step-up increase variables for slope instabilization.

= Gradual increase variables for slope instabilization. According to this equation, the geomorphological process from the forming of the slope

to the occurrence of the landslide is explained

as follpws (Fig. 2, reference 12):

a. A slope of V-shaped dissected valley

(formed tens of thousands of years ago) in Pleistocene polygenetic volcano was overlaid by aeollan unconsolidated thin pumice layer. This

layer was overlaid by alternating beds of lava

and pyroclastics to develop a piled-up long slope. Subsequently, a lava layer covered the

foot such as a "weighting banking."

b. With a decrease in the supply of debris

from the slope due to downward tendency in temperature and ascension of forest line in Pleistocene (from ca. 10,000 years ago), downward erosion of valleys (50 m in relative

height traced back from a geomorphic surface

downward) and retreat of the knick point

(falls, for instance), the slope was dug at the foot and became unstable (expression V„ and

Vq as in the above equation) . In the front

of the V-shaped buried valley where "weighting

banking" was removed by this erosion, a large-scale landslide had previously occurred

caused by heavy rainfall and/ or large earthquakes (Vp) . The former large-scale

landslide valley was formed in this process.

c. Later, the falls retrograded about 400 m, the valley bottom at the foot of the landslide

site rapidly lowered its level. This caused deterioration of the slope of the mountain

where the buried valley was hidden and intermittent creeps (estimated from existence

of slid scarplets, etc.) advanced (V„ and

d. A large-scale landslide occurred from Vp , preceding rainfall of 160 mm and an earthquake with seismic Intensity around VI. Groundwater catchment and storage were high because it underlaid a buried valley. The earthquake amplification (acceleration) was large because of a piled-up ridge (see the A factor in the above equation).

In this case, the direct factor Vp (final forcing variable for generation of landslide)

and A (amplification coefficient for Vp

by site condition) were large and played an important role for the occurrence of this landslide; but they alone did not cause it. It is important to realize that the Instability

Index I first exceeded the critical value when

V„ and A were added to the precedent facts

that the level of the valley bottom already had been lowered and the slope had deteriorated;

thus, the landslide. From this viewpoint, £

and b above can be explained rationally. In addition, the viewpoint of the historical evolution (development) of slope landforms makes it possible to predict landslide site potential and reasonably explain the mechanism

of occurrence of a landslide.

3. GEOGEIAPHIC WORKS OF GEOGRAPHICAL SURVEY

INSTITUTE

3.1 Earthquake Disaster and Landform Condition Earthquake disasters are closely connected with landform condition. Disasters caused by earthquake movement and liquefaction are not directly affected by landform condition but by underground conditions. Reconstructing

paleogeography by geomorphological methods and researching the historical evolution of landform provides a good understanding of underground condition; ground condition is determined by geomorphological place at the time of sedimentation (= landform formation) . The ground condition of surface layer, not deeper than several meters, is closely related

to mlcrolandform (references 13 and 14). Therefore, it is possible to determine ground condition by reading the mlcrolandform. It is possible to determine ground condition, such as evaluation of underground, by reading a Landform Classification Map. It is recommended

to use this map from the viewpoint of earthquake disaster prevention. Fig. 3 shows a model of the relation between geographic place and sedimentary environment ground condition.

Fig. 4 shows a sea level change since the last

265
glacial age which had a great influence on the sedimentary environment of the coastal plain. The Geological Survey Institute published a Land Condition Map with landform classifications. The outline of this map is explained

in Section 3.3. Earthquake damages include not only direct damage of material and life but also overall damage to living and productive activity. It is necessary to collect various kinds of geographical data to predict such damages. For this purpose, the Geographical Survey

Institute's Land Use Map and digital

geographical data (Digital National Land Information) are available. These are explained in the following paragraphs.

3.2 Land Use Map

The map is categorized into three major

divisions: urban area, agricultural land and forest. The urban area is subdivided into intermediate divisions and/or small divisions.

Here, the regional planning division is used

for Town Planning and Zoning Acts, and urban function is given special attention. The agricultural land and forest are subdivided

into intermediate and/or small divisions,

paying attention to the vegetations. Six different colors are used on the maps. By combining the shades of color and patterns,

this map differentiates 35 land use divisions. Maps denoting land use (Fig. 5) are based on interpretations of aerial photographs by surveys and other data.

3.3 Land Condition Map

Before and after Typhoon 5919 (Isewan Typhoon,

1959), surveys on landform and flood disaster

were performed in Nobi Plain. These surveys revealed that is was possible to predict the location and intensity of damage by characteristics and historical evolution (development)

of landform, altitude, and history of reclamation. This experience accelerated the development of the Land Condition Map that focused on landform classification in lowlands. Investigations were then performed when earthquakes occurred. The results showed correlations between micro- landform ground conditions and earthquake disaster. Therefore,

this type of landform classification map was considered useful.

Later, topographical classification and dip division of mountains were added to the map to understand slope disaster potentials. Since

1980, the information on underground features

has improved. Contour lines of the base of alluvium deposits (= ground surface at last glacial stage) or isopach maps of soft layers were Included as well as topographical

classification.

The contents of the Land Condition Map include landform classification, ground elevation, and various organization and public facilities pertaining to land disaster and development.

As seen in Fig. 6, maps were made mainly on the basis of the interpretation of aerial photographs and field survey. Units of the landform classification are classified into groups homogenous in morphological unit, morphogenesis (natural processes such as erosion by running water and sedimentation,

volcanism, crustal movement, etc.), and formative age. Mountains and hill slopes are divided by slope and subdivided by shape of slope. Terrace and tableland are subdivided by formative age and elevation. Lowland is divided into general surface, relatively higher

land, and well drained landform, and furthermore, subdivided by microtopography. An artificially deformed area, with bad conditions, is denoted so the former units of landform classification, before artificial

deformation, can be shown. In addition to the above, land is contoured every 1 m to express subtle undulations on the ground surface.

3.4

Digital National Land Information

Information from maps such as the Land Use Map

and the Land Condition Map is not always readily available as in the case of obtaining

quantitative data from measuring. The Geographical Survey Institute initiated supplying various geographic digital data accurately corresponding to 1:25,000 topographical map. Now data, such as digital

data, is more easily obtained.

The Digital National Land Information is stored

on magnetic tapes. A large quantity of data can be processed and analyzed very fast by an electric computer. The data is stored as line data, e.g., data about shoreline, river channel, the administrative sector, roads and railroads, etc., and data as mesh (about I km^)

4.0 SUMMARY AND FUTURE VIEW Ground disaster accompanied by an earthquake is greatly affected by underground

characteristics. The ongoing land condition survey by the Geographical Survey Institute is focused to know where earthquake disaster potentials lie. The Land Condition Map contains much useful information.

It was concluded that work needs to focus on the response characteristics of ground by computer analysis and liquefaction potential.

Mappings on such were carried out. They were reported and discussed at UJNR meetings. There

are "distribution map of ground motion" and "map of liquefaction potential" (mesh map) by Tokyo Metropolis (reference 15) as an example

of mapping a large area. This mapping.

266
however, is based on the assumption that much

borehole data can be obtained in urban areas in Japan.

It is necessary to study the landform classification of landforms in a tropical humid

climate and an arid climate. Such landform classification maps at such places are expected

to increase. Furthermore, to inform map readers about the data contained on the maps, it will be necessary to produce not only general landform classification maps but also maps that meet various specified purposes such as an earthquake disaster hazard map and flood hazard

map, etc. In that case, synthetic applied

geomorphological evaluation is needed (reference 16). It is necessary to evaluation

new characteristics of combined units Including

the characteristic of each landform classification unit. On the other hand, it is necessary to assume even changes of the original characteristic when natural and social

(artificial) conditions are added to landform

condition. Landform classification map

expressions must be simplified by using colors

and patterns to help map readers understand the similarity and contrast in the units of landform classification.

5. REFERENCES

(1) Ogasawara, Y., "Earthquake Damage and Disastrophism,” Journal GSI, Special Ser. 2,

13 p, 1949 (J).

(2) Ohya, M., "Chilean Earthquake Tsunami and Coastal Landform (I)," Journal GSI, No. 26,

p 6-13, 1961 (J).

(3) Ohya, M., "Chilean Earthquake Tsunami and Coastal Landform (II)," Journal GSI, No. 27,

p 12-19, 1962 (J).

(4) Geographical Survey Institute, "Report on

the Niigata Earthquake," 60 p, 1965 (J).

(5) Kobayashi, M. and Magome, H., "Disaster

of Niigata Earthquake of 1964 and Land Condition Survey;" Journal GSI, No. 30, p 1-7. 1965, (J).

(6) Takasaki, M., et al., "The Reration of the Damages Caused by the Niigata Earthquake

and the Topographical Conditions in Niigata

Plain," Rep. Cooperative Res. Disaster

Prevention, No. 11, p 13-18, 1966 (J).

(7) Geographical Survey Institute, "Report on the Izu-Oshima-Kinkai Earthquake of 1978,"

56 p, 1979 (J).

(8) Kobayashi, M., "Disastrous Ground Failure

in a Residential Area of Large-Scale

Cut-and-Fill in the Sendai Region Caused by the

Earthquake of 1978," Bulletin of Geogr . Survey

Institute , No. 24, p 15-25, 1980 (E).

(9) Tsurumi, E., et al . , "Relationship Between Seismic Disaster and Ground Condition

in Noshiro Area, Akita Prefecture," reported at the 16th Joint UJNR Meeting , 1984 (J,E).

(10) Nagaoka, M. , et al., "Geomorphological Characteristics and Factors of the Large-Scale Landslide and Detritus Flow on the South Slope

of Ontake Volcano Caused by Naganoken-Seibu

Earthquake, 1984," reported at the 17 Joint UJNR Meeting . 1985a (J,E).

(11) Hatano, S. and Nagaoka M, "Geomorphic

Properties and Causal Factors of the 1984 Ontake Landslide, Analyzed by Time-Series

Air-Photos," Bull. Vole. Soc. Japan, Ser 2, Vol 30, p 107-108, 1985 (J). (12) Nagaoka M. et al., "Casual Factors of Large-Scale Landslide as a Viewpoint of Postglacial Historical Evoaluation of Landforms

- on the Case of Naganoken-Seibu Earthquake,

1984," Abstract of the 15 Meeting of Assoc. Quart. Res. Japan , No. 15, p 90-91, 1985b (J).

(13) Tsurumi, E. et al., "Seismic Disasters

and Land Conditions of Lowland," reported at the 14th Joint UJNR Meeting , 1982 (J,E).

(14) Sunaga, H. and Kumaki, Y., "Earthquake Disaster and Landform Classification, Journal GSI , No. 56, p 24-31, 1982 (J).

(15) Tokyo Metropolis, "Report on the Prevention of Earthquake Damages in the Tama-Region," 576 p, 7 maps, Tokyo Met. Gov., 1985 (J). (16) Kanabubo, T, "Present Situation and View

of Landform Classification," Journal GSI , No. 56, p 5-6, 1982 (J). (17) Inokuchi, T., "Large-Scale Landslide and Rock Avalanches at the Head Source of the Denjo River," Natural Disaster Research Studies,

No. 25, p 45-62, National Research Center for Disaster Prevention, 1985 (J). (18) Mizutani, T. , "Disaster Prevention and Landform," Kokon-Syoin, 168 p, 1982 (J). (19) Ikeda, T., "Study on the Alluvial

Deposits on Tokaido Region," Contribution

Institute Geol. Paleontol., Tohoku University,

No. 60, p 1-85, 1964 (J+E).

( ) Kadomura, H., "Systematic Aerial Photo Analysis of Soft Ground Condition," Geogr. Rev. Japan, Vol 41, p 19-38, 1968 (J+E).

( ) Kobayashi, M. et al., "Investigation on

the Landform Condition and Earthquake

Disaster," Report on the Mlyagi-Ken-Oki

267
Earthquake of 1978, Sci. Tech. Agency,

p 131-173, 1980 (J). Note;

(J) Written in Japanese only.

(J+E) Written in Japanese with English

abstract

(E) Written in English only.

(J,E) Written in English and Japanese.

Fig. 1 Stereoscopic topography of the 1984 Ontake Landslide (reference 17).

Fig. 2 Projected profile of 1984 Ontake Landslide (revised from reference 12).

268
fan natural levee 2one delta (or sand bank. dune)

3-b. Schematic geologic section of alluvial

(postglacial) deposits filling the drowned valley.

sand bank tlulval deposita

(a) Alluvial (postglacial) deposits of a big river flowing into a bay or a small

r iver

.

fan

(b) Alluvial (postglacial) deposits of a big river whose alluvial fan directly faces the sea.

3-d. Schematic geologic columns of alluvial (postglacial)

deposits in the lowland (compiled by the Geological Survey Institute from reference 19)

.

Fig. 3 Relationships of underground condition and landf orm-sediment

environment of fluvial lowland (reference 19)

.

269
ig • ^ Sea level change since the reference 18 from Kaizuka,

last Glacial Age

1976) . (simplified by

Planning Base map making

Preparing the photos

Preparing the reference dau

Photo-interpretation Analyzing the data Appearance of the planned area Checking the revised parts of the base map

Source material collection in the public bodies

Filling the source material

Observation of the field Cheeking the preparatory work carried out Confirming the changed land use after photographing Checking the map source material collected

Filing the dau observed in the field Inking the data on the photos Arrangement of data on the overlay Adling the reference maps

Revision of the base map

Land use classification by photo-interpretation Compiling the revision data map for drafting

Tie-strip making

Marginal information making

(Survey dau: compiled manuscript)

Revising the positive sheet of the base map

Scribing the boundary sheet Mask-sheet making Making the compounded positive and negative sheets (Drafting dau; compounded positives and negatives for plating)

Plating and reproducing

(Final daU: Printed Land Use Map)

Fi£,. 5 Work flow of the 1:25,000 Land Use Wap.

270
I Drawlin «nd Mip-H«productloa

|

Fig. 6 Work flow of the 1:25,000 Land Condition Map.

271

THEME m

Storm Surge and Tsunamis

SATELLITE METHODS AND APPLICATIONS

IN MARINE ENVIRONMENTAL ANALYSIS

C. S. Barrientos, F. G. Everdale and R. P. Stumpf

Marine Environmental Assessment Division Assessment and Information Services Center National Environmental Satellite, Data and Information Service

National Oceanic and Atmospheric Administration

Washington, DC 20009 USA

ABSTRACT

Satellite can provide important data

for the study and monitoring of processes, conditions, and changes in the marine environment. Remote

sensing data have good spatial and

temporal resolution, thus providing

adequate information for analysis at reasonable cost. Three

characteristics can be determined

directly on water by satellite

measurements, 1) surface temperature,

2) turbidity; and 3) water color. Each of these characteristics can be

used to identify certain water masses

or water parcels to estimate the surface concentrations of sediment and

pigments. Overland, changes can be

seen by satellite from images taken at different times. Similarity, changes

in the boundary between land and water

can be studied from time series

imageries.

Satellite data were used together with

conventional observations and modeling

to study the Chesapeake Bay and the adjecent shelf waters. It was also

used to study the warming episode in the eastern Pacific Ocean, to assess

the fishery situation in the

Philippines, and to determine the feasibility of satellite applications

to fishery off West African countries.

1. ItfTPOPUCTXQN

Satellite can provide important data

for the study and monitoring of processes, conditions, and changes in the marine environment because of their synoptic and repetitive

coverage. Conventional oceanographic

data, such as that collected from isolated moorings or from individual

cruises, have substantial spatial or temporal limitations. Synoptic data

can be collected only at great

expense. Spatial variability cannot

be properly observed from a limited

number of discrete stations or tracklines and cruises cannot often be

repeated. Satellites, however, collect data at thousands of points

over the water and adjacent land thereby providing essentially

continuous coverage. Certain sensors, such as those on the NOAA weather

satellites, provided coverage as often

as two to four times a day, permitting

the modeling of dynamic estuarine and coastal processes. However, they cannot provide infomation on the vertical structure of the water

column; therefore, a combination of in situ measurements, remotely sensed

data, and modeling results are needed

in the study of estuarine and oceanic

processes.

At present, data obtained from satellites yield information on temperature, turbidity, and water

color. Each of these characteristics

can be used to identify certain water

masses or water parcels and to estimate the surface concentrations of sediment and pigments. Therefore,

satellites may aid in monitoring

fronts and plumes and in identifying

circulation patterns in Cheaspeake Bay and on the adjacent shelf. There are three satellites sensors of potential use in estuarine waters:

the NOAA-n series Advanced Very High

Resolution Radiometer (AVHRR)

.

Nimbus-7 Coastal Zone Color Scanner

(CZCS) , and Landsat Multispectral

Scanner (MSS) (Tables 1 and 2) . These

sensors are on polar-orbiting

satellites; i.e., the satellites have

orbits oriented north-south in order

to allow coverage of the entire globe. The sensors measure the radiance of the reflected visible and near- infrared light and the thermal

infrared energy emitted by the earth

and atmosphere. Each sensor system

has different characteristics, which

provide different advantages.

275
Landsat was designed for use over

land, which has greater and more variable reflectance than does water. Accordingly, Landsat data has been

shown to provide a good measure of suspended solids in turbid water. However, the satellite sensor is insensitive to slight changes in the reflectance of clear water. Landsat

also has very high resolution, 80 m. To achieve this high resolution, the satellite has an overpass period of 16 days, thus limiting the temporal

resolution.

The CZCS samples every three to five days in the mid-Atlantic with a resolution of 800 m. It has five reflected light bands and one thermalIR band, all specifically designed for oceanic waters. It is quite sensitive

to variations in water brightness,

but it is not always usable for turbid

waters, like those found in the upper

Chesapeake Bay, that reflect enough

radiance to saturate some channels on the radiometer. Algorithms for the

CZCS data have produced" estimates of oceanic chlorophyll content accurate

to within 30 percent (Gordon et al.,

1983)

. These results have permitted

analysis of biomass on the shelf and

in the Gulf Stream, and detection of some eddies and other features of circulation. Unfortunately, the error

in these algorithms increases with

turbidity, making them highly

unreliable in many estuaries.

Nonetheless, because the sensor's

spectral bands were designed to collect measurements over water, this

satellite sensor can provide useful information on estuaries. The CZCS is eight years old, well beyond its expected lifespan, and has failed in June 1986. Thus, we can no longer

depend on it as future source of data. However, data collected during the eight years of operation are very

useful in retrospective analysis.

The AVHRR sensor is on the NOAA-n

polar orbiting satellites series. It measures radiance in two reflected

(visible and near-infrared) bands and

two or three thermal infrared bands. NOAA has developed regression

equations to calculate sea surface temperature (SST) using data from the thermal channels (Strong and McClain,

1984)

. The sensors are intermediate

in sensitivity between those of

Landsat and CZCS, making them useful

in estuaries (Gagliardini et al.,

1984) . This satellite is designed for frequent sampling of the globe, twice

per day with one satellite, and with

two operational satellites, four overpasses (two daytime and two

nighttime) may be recorded. The best resolution of the AVHRR is 1.1 km, making it useful for delineating the larger estuarine and oceanic features. For application to Chesapeake Bay, the AVHRR and the CZCS systems are preferable because either sensor can include the entire Bay in one scene, and because either system has a high

enough sampling frequency to aid in analyzing coastal dynamics and in verifying models. Biweekly sampling,

like that done by Landsat, supplies

more limited information on the dynamics of an estuary.

2. TEMPERATVRa. ANALYSIS

Using Planck's Law, we can calculate a temperature from the thermal infrared

radiation detected by the satellite

sensor. Due to atmospheric absorption

and emission (primarily by water

vapor) , the thermal radiance reaching

the satellite is not the same as that emitted by the surface of the water. Often the calculated black body

temperature is lower than the surface

temperature. A technique for atmospheric correction and calibration

is required to determine the true surface temperature from the satellite

data. The most accurate atmospheric

correction entails the use of two or more thermal channels. Because CZCS

has only one thermal-IR band,

atmospheric corrections for CZCS are

less accurate, although we can obtain

reliable data on temperature

gradients.

The AVHRR, through the use of regression equations developed at NOAA

(Strong and McClain, 1984) , provides excellent estimates of sea surface temperature using channels 3 and 4, and also channel 5, when available. These calculated, "multi-channel sea surface temperatures" (MCSST) , are accurate to within 1 °C when compared

to ship and buoy data. Much of this 1 °C diffence probably results from differences in the measured property ; buoys measure bulk temperatures of the

276
upper 0.5-1 m of water, whereas

satellites receive radiation from the

top millimeter or less of the water

(skin temperature) . Errors can also

result from incomplete correction from

the atmospheric effects, particularly

for the effects of dust, fog, or thick

haze. Because NOAA-6's channel 3 thermal data were noisy prior to September 1982, and the satellite did

not have channel 5 (found only on NOAA-7 and NOAA-9) , MCSST equations

were not available for the April 1982 NOAA-6 imagery. Therefore, in this

study, sea surface temperatures for

the NOAA-6 imagery were determined by

linear regression of channel 4 brightness temperatures against MCSST

calculated from NOAA-7 data from the

same day. The NOAA-6 temperatures

matched the NOAA-7 temperatures to within 1 °C (95 percent confidence

level) . 3. TTIRBIDITY DETERMINATION

Suspended materials provide useful

tags of water parcels, and they

indicate the extent of runoff, land

erosion, and nutrient discharge.

Visible or reflected infrared light

provides an indication of the quantity

of suspended materials by showing the turbidity of the water. However,

monitoring and comparing imagery from different days requires a consistent

and quantifiable property such as reflectance, or suspended sediment

concentration. Determination of these

quantities from the raw data requires

several corrections of the total measured brightness.

Processing an image first requires the elimination of areas covered by clouds

or showing discernible sunglint.

Sunglint is the specular (mirror-like)

reflection of sunlight off the surface

of the water. It may provide useful information on surface phenomena, such

as wave fields, but it prevents us

from obtaining information on materials in the water, by masking the water column reflectances. The effect

is similar to trying to see out a window at night and seeing only your

own reflection. Water is a highly

effective absorber of near-infrared

radiation, therefore, both clouds and sunglint can usually be detected

through the increased brightness they

produce in near-infrared bands. In

the areas not containing clouds or sunglint we can obtain reflectance of the water column. To obtain reflectance in these areas, we must

correct for both the radiance of the atmosphere between the earth and the satellite—known as path radiance--and

the amount of total radiance reaching

the surface (the incident irradiance)

.

Over water, path radiance frequently

constitutes 50-90 percent of the total visible radiance detected by the satellite. It changes with the dust

and water vapor content, and therefore

can vary from one image to the next.

In addition, the total path radiance

generally increases toward the limbs

of the image because of the greater

thickness of the atmosphere. To obtain only the radiance leaving the

water, we must substract the path

radiance from the total radiance

observed at each pixel by the satellite. The path radiance is usually determined as the radiance observed from the darkest pixels in the locale. In estuarine work, the path radiance is assumed uniform over

the study area—a functional, albeit

not ideal, correction. Areas near the

limbs of the images are generally not

usable. The amount of radiance leaving the water depends on the amount entering;

hence, there is need to correct for the incident irradiance. The incident

radiance depends on the sun's

elevation. Thus, water with a given

sediment content will appear darker in the morning than at noon, and darker

in the winter than in the summer. In contrast to the atmospheric

correction, a calculation of the sun's elevation can be made quite precisely

throughout the image, thereby allowing

an estimate of the incident

irradiance, using the solar constant multiplied by the sine of the solar elevation (Gordon et al., 1983). By

dividing the atmospherically corrected

water radiance by the incident

irradiance, we obtain a form of reflectance that can be compared from

one scene to the next with the same sensor band. We studied early morning

and early afternoon scenes for the same day on 14 April 1982. Without

the correction to obtain reflectance,

the morning scene would show about 60 percent of the turbidity of the

277
afternoon scene. With the corrections, the turbidity becomes comparable in both scenes.

In the estuarine surface waters, reflectance (R) tends to vary directly

with the common logarithm of the sediment concentration (n^)

R <proportional> logio(ns)

so that a determination of reflectance

can also show the amount and variability of suspended materials in the Bay (Munday and Alfoldi, 1979). For the study period, we do not have

data on the concentrations or suspended solids, hence the results

are shown as reflectance and not as concentrations. However, comparisons

of reflectance (corrected turbidity)

on different days can help show distribution patterns and paths of transport of suspended materials. In modeling, the suspended material may

itself be of interest, or it may serve

as a surrogate for other substances or organisms.

4. PIGMENT ESTIMATION

As described earlier, algorithms

developed for the CZCS have given very

good estimates of chlorophyll in oceanic and shelf waters. However, the techniques do not provide valid

information in many estuaries, because

the water's turbidity interferes with

the atmospheric correction (Gordon et

al., 1983). An algorithm developed in our Center in NOAA does permit the estimation of relative variations of the chlorophyll concentrations and the detection of algal blooms in turbid water such as found in Chesapeake Bay

(Stumpf and Tyler, 1986) . The technique is applicable to both AVHRR

and CZCS data, although the CZCS,

owing to its narrower bands and greater sensitivity, would tend to give results that are more specific to chlorophyll-a. The AVHRR, having

broader bands, would respond to variations in other pigments, although

it would be principally sensitive to chlorophyll. The corrected radiance

as described in the previous section

is necessary in these calculations.

the analyses can be displayed

meaningfully only in color and not included here. The images showed that blooms appear in the western

tributaries of Chesapeake Bay, the Rappahanock, Potomac, Patuxent rivers. The upper Bay, in the region of the turbidity maximum, contains low pigment concentrations. The frequent overpasses of the sensors permit study

of diurnal variations in phytoplankton

blooms. Many of the blooms result

from algae transported into the area, so that modeling the drift would be helpful in predicting the onset of these blooms. Conversely, detection

of blooms with satellites can provide

data to aid the calibration of drift

models.

5. TTIftCrf- PT^EPARATIQM The AVHRR and CZCS collect the data by

scanning in a line across the satellite's path, collecting about

2000 radiance measurements on each

line. Each measurement is called a pixel (picture element) ; the area of the earth viewed in each pixel

determines the resolution, 1.1 km for AVHRR and 0.8 km for CZCS. As the satellite continues in its orbit, each scanline is slightly offset from the preceding one, thus an image results

from the accumulation of these

scanlines. The satellites take 1-2 minutes to produce an image of Chesapeake Bay. An image differs from

a photograph, wherein all points are recorded simultaneously.

The original images contain

distortions that are caused by the earth's curvature and the viewing

angle of the satellite. To correct

for the distortion, all the images were geometrically stretched to match

certain points in the image with

reference points on a Mercator

projection. Because this commonly

used techinque does not assure correspondence with a Mercator

projection throughout the image, we call the result a pseudo-Mercator

projection. On the resultant image

the AVHRR data have a 1.2 km/pixel

projection, and the CZCS data have a 0.9 km/pixel scale. We applied our analysis techniques for

the pigments for April 1982 for AVHRR

and CZCS imageries. Unfortunately,

278

A total of 19 images were used in the Chesapeake Bay study (Table 3) . We
included the Table here so that future

users may recognized that these images

have already been analyzed. Land and clouds were masked in the images using

the near-infrared band (AVHRR channel

2, CZCS channel 5). Water absorbs near-infrared light quite effectively;

therefore, a brightness threshold can generally be used to distinguish land

and clouds from water. Highly turbid

water may have the same reflectance as

land, therefore some tributaries may

be partially masked.

The temperature images were colorcontoured in 1 °C intervals between 7 and 18 °C, where (for reference) the 7 °C contour contains the data from 6.95

to 7.95 °C. Temperatures outside that

range were not distinguished.

The reflectance (turbidity) data were

grouped into 16 classes and colored.

Small values of reflectance should be interpreted as water that is relatively clear, i.e., the water does

not contain large quantities of material, either organic or inorganic

in suspension. Larger values of reflectance denote increasing

quantities of suspended material in the upper water column. Reflectance

indicates only the degree of water

clarity or the relative quantity of materials in the water and does not indicate the composition of those

materials.

6. SATELLITE IMAGERY ANALYSIS

AND INTERPRETATION

The distributions of temperature and turbidity (reflectance) in the

satellite imagery are used to infer

some of the circulation dynamics and water quality of the Bay and the adjacent shelf and slope region. The

region contains four principal water

masses: the Bay waters, the shelf

water, the slope water, and the Gulf

Stream. Temperature imagery for the study period can be used to discern

variations in all four. Reflectance,

however, provides information only on the Bay waters; the sensitivity of the sensor and the quality of the atmospheric correction cannot discern

variations in the reflectance of the comparatively clear offshore waters

during this period.

7. ANALYSIS FOR CHESAPEAKE BAY Insolation can significantly influence

the surface skin temperature of the

Bay. In the satellite imagery, the

Bay cools by 1 to 4 °C at night, and warms throughout the day. Reduced

vertical mixing during periods of reduced wind speeds may promote this

variability. Satellite image for April 12, 1982 shows the low nighttime temperatures in the Bay. The surface temperatures fall below 8 °C

in the upper Bay and below 7 °C in the lower Bay (below the Rappahannock

River) . By early afternoon, (shown by the afternoon image) , diurnal heating

has greatly increased the temperature.

The entire Bay has warmed by as much

as 4 °C. In situ data taken in the

Bay on this day correspond to the satellite-derived daytime

temperatures. The shelf waters have warmed to a lesser extent (1 to 2 °C)

,

while negligible changes appear in the slope and Gulf Stream waters. The difference between the day and night

temperatures observed during the study

period may result from either diurnal

variations in the skin temperature or

fog and mist over the Bay at night.

If the variation is in the skin

temperature, it does not necessarily

indicate temperature fluctuations in the upper meter of the water column, because the skin temperature can react rapidly to changes in air temperature

and sunlight.

The seasonal increase in surface temperature is evident during the study period. Some warming is evident

on 15 April image analysis. By 23 April, the Bay and shelf have warmed

substantially, about 4 °C throughout

the area. Shallow waters, such as those in the sounds and tributaries,

respond relatively quickly to changing

insolation and air temperatures, while

the deeper waters of the Bay and shelf respond more slowly. Thus, the western tributaries and eastern shore

of the Bay contain warmer water than

does the central Bay. Similarly, the lower Bay has warmer water than the adjacent shelf. Therefore,

temperature can be used to infer transport of tributary water in the

Bay and certain characteristics of the exchange of water between the Bay and

the shelf.

279
The flows from the four largest

tributaries, the Susquehanna, Potomac, Rappahannock, and James appear to hug

the western shore. This is particularly evident on 12 April image

and other data, where cool northern

Bay and Susquehanna water follows the western shore, as does the 9 °C water

leaving the Potomac River. The warm

waters associated with the Rappahannock, York, and James Rivers

appear to flow along the western shore down the Bay and out the mouth around

Cape Henry. The reflectance is generally highest

in the upper reaches of the Bay and

its tributaries, decreasing with

distance downstream. The main Bay, Potomac River, and James River show turbidity maxima at the upper reaches with a rapid decrease in reflectance

and fairly clear water downstream, as material supplied by the rivers

settles from the surface. In the main

Bay, lateral variations appear in the turbidity maximum.

The water of the central Bay, off the Potomac and Rappahannock Rivers is clearer and colder than the adjacent

waters. These differences result from

the greater depth of the water, which

reduces both warming and sediment

resuspension, and the greater distance

from riverine and shoreline sources of sediment. The thermal and turbidity fronts at the mouths of the Rappahannock, York, and James Rivers coincide. The Bay mouth plume shows a more complicated

structure. The temperature and turbidity plumes do not always

coincide, although the turbidity

plume, owing to its low reflectance,

is less pronounced than the plumes and

fronts within the Bay. The thermal plume's position is highly

variable. Early on 12 April, the warmer Bay water appears to branch;

one limb hugs the coast to the south

for some 100 km, the other limb moves directly offshore, suggesting intense

mixing. This structure may result

from the 12 kn winds from the west

pushing water offshore. In the afternoon the plume is oval, although

tending southward. A bulge or turbid water appears off the mouth of the Bay

at this time. A thermal plume is

distinct, despite the maximum flood tide at the mouth of the Bay. Small detached pools of 10 °C water are observed offshore. Two or three elongated filaments suggest intense

mixing. Again, as on 12 April, the plume extends to the south. The 10 °C ribbon of water along the North

Carolina cost may be simply warm

nearshore water and not water from the

Bay.

8. RESULTS FOR OFFSHORE WATERS

The water masses of the shelf, slope, and Gulf Stream show strong mixing and marked spatial variability in temperature during the period. The shelf waters are cooler (3 to 11 C) than the other water masses. Diurnal

fluctuations in observed surface temperature occur in the shelf waters, but to a lesser degree than in the

Bay. Near Cape Hatteras, the warm

Gulf Stream (>15 °C) pinches off the shelf waters of the mid-Atlantic

Bight. The Gulf Stream veers to the northeast from Cape Hatteras. Because

the slope water originates as a mixture of shelf and Gulf Stream

waters, it has intermediate

temperatures to shelf and Gulf Stream

waters. Thermal fronts sharply

delineate the three water masses: the shelf break front between the slope

(blue-green) and shelf (purple) waters, and the north wall of the Gulf

Stream. Gulf Stream eddies also can be tracked

with the imagery. A warm core eddy,

Eddy 12, having clockwise rotation,

lies about 180 km off the mouth of the

Bay. The eddy becomes increasingly

evident through the study period

because a filament of warmer Gulf Stream water wraps around it. The filament of 15 °C water develops on the SW side of the eddy (the eddy center is green) . On the NE side, cold shelf water is apparently being

advected onto the slope by the eddy.

By 15 April, the filament has partially encircled the eddy and the cold water (10 °C) has moved around

the eddy almost to the Gulf Stream.

On 23 April, the filament has encircled the eddy and the cold water

has moved between the eddy and the Gulf Stream. The eddy also appears in the chlorophyll pigment image. As the Gulf Stream has much lower chlorophyll

280
than the slope or eddy, the filament

appears as a ring of lower chlorophyll

water.

9.

WARMING EPISODE IN EASTERN

PACIFIC OCEAN

Periodically in the Pacific Ocean, there is a large global episode of abnormal meteorological and oceanic

reversals of the circulation patterns

causing some major environmental

events worldwide. Some of the manifestations are good and some are

disastrous. This event is generally

referred to as "El Nino" Southern

Oscillation (ENSO) episode. The last major ENSO was in 1982-83 which caused

damages in the billions of dollars. Normally in the Pacific Ocean, the

trade winds from both hemispheres blow

towards the west bringing warm water

in the Western Pacific (Australia,

Indonesia, and the Philippines) , thus keeping the western Pacific warm and

the eastern Pacific relatively colder. Part of this normal pattern is the presence of cold upwelling region off

the coasts of Chile, Peru, and

Ecuador. For reasons still unknown,

this normal pattern is reversed every

so often in which case we refer to as the ENSO episode. The strongest

indication of ENSO is the warming in the eastern Pacific off the coast of South America. During the 1982-83

ENSO episode, we looked at the archive

for satellite data for analysis of the

event. It was during this event when

we initiated our study of satellite

applications for the eastern Pacific

Ocean. From October 1985 to May 1986, we requested that the NOAA Polar orbiter

satellites collect AVHRR data for the coast of South America in Eastern

Pacific region. The AVHRR data were

of Local Area Coverage (LAC) type

which has a resolution of about 1.1

km. Every two or three weeks during

the period, we were able to obtain a series of two to four days of relatively clear imageries. The daily

images were analyzed into a Mercator

Projection, then composited to eliminate some of the clouds. Series

of Sea Surface Temperature (SST) maps

were produced which show the evolution

of warming if there is any. During

early 1986, several groups have issued

advisories of possible ENSO episode. However, our analysis showed that this was probably just the localized "El Nino" and not the large scale ENSO. We were able to reach that conclusion

because of the better resolution (LAC,

I. 1 km) data that we used.

10.

ASSESS THE FISHERY SITUATION

IN THE PHILIPPINES

Fishing industry is a major part of the livelihood in the Philippines.

Majority of the population depends on fishing for their subsistence and the industry itself is a major component

of the country's economy. It is important therefore to know the fishery situation around the Philippine islands. Remote sensing data is ideal to determine the major oceanic features around the Philippines. By collecing

AVHRR LAC imageries, we can determine

the locations of warm and cold

regions, and sharp temperature

boundaries. These oceanic features

are related to productivity and food abundance which indicates better

fishing. We have coordinated our earlier efforts with the Philippine

government officials and they have concluded that the correlation of the oceanic features with their fishing

data (catch statistics and larval

surveys) is very informative and

useful. The information can provide

them the understanding of the fishery

stocks and assist them to save fuel during actual fishing voyage. We are continuing a series of analysis of imageries for the area. The main

problem for this application in the Philippines is that very few times

during the summer that we can find clear image.

II. APPLICATION OF SATELLITE IMAGERY

The National Oceanic and Atmopsheric

Administration entered into an agreement with the U.S. Agency for International Development (AID) , to study the feasibility of applying

satellite imagery to the management of west African fisheries.

The geography, meteorology, and oceanography of the West African

281
coastal region provide for favorable conditions for strong upwellings. The relatively broad and shallow west African shelf has a well-defined

seasonal upwelling which is easilly

identified in satellite images. The upwelling cells extend to 80 km off

the coast. The region of upwelling

extends from Cape Bojador to below

Cape Vert. The most intense upwelling

occurs in the spring and is responsible to local winds along the

coast. We collected and analyzed satellite

imagery for January through September

1983 to determined the feasibility of analysis of upwelling or other

features of the environment using NOAA AVHRR data. We compared the observational characteristics of the several satellite systems available

for adequacy of coverage and resolution. We found the Meteosat

(European) was the best for temporal coverage (hourly) while the NOAA AVHRR

was the best for spatial resolution

(1.1 km). The French are presently

providing basic maps of weekly sea surface temperatures of the region

based on Meteosat imagery. The West African fishery is dependent

on the upwelling cells for nutrition

and possibly life-cycle timing for reproduction. Each upwelling cell ages after being brough to the

surface. The water is advected by surface flow, warmed by the sun, and

has its nutrient contents depeleted by

the grazing zooplankton. The fish in turn feed on the abundance of plankton. Observations indicate the

local abundance of fish north of Dakar may be predicted somewhat by the earlier size and intensity of upwelling cells there, as the fish appear to migrate from one aging

upwelling cell to a newer one in a slow southward movement. Because

fisheries of the region are strongly

linked to the upwelling cells, and because the upwelling intensity is highly variable, timely satellite

images showing the intensity and location of these cells is valuable to fishermen and fisheries managers. The

next task is to develop some kind of predictive indeces or measures of these indicators that the fishermen

and managers concern can use.

12 . SUMMARY

Data collected by satellites can show spatial and temporal changes in temperature, reflectance, and color

(pigments) in the Bay and offshore.

The high sampling frequency of the NOAA-n series and CZCS permits

detection of daily and weekly

variations in temperature.

Temperature contrasts during the period were strong, the imagery

showing fronts of 4-5 °C and a temperature range of 15 °C. Although

turbidity showed slight changes over

the study period, spatial variability

in turbidity is quite evident in the

images. The images show the turbidity

maxima near the heads of the Bay and

the major tributaries, and the clear water in the lower central Bay and offshore.

The Bay mouth temperature and turbidity plumes were neither well defined nor (apparently) persistent,

thus movement during the period could

not be inferred from a set of images. The greater temperature contrast

offshore and the persistence of distinct features such as Eddy 12 did allow tracking of these features by

satellite. This indicates that the movement of comparably distinct

features within the Bay (such those produced by floods) could be traced. Analysis of the imagery allows us to infer characteristics of the Bay's surface circulation for the conditions

of the study period. The Susquehanna

River waters produce a southward flow along the western shore, probably

because of Coriolis force. As the Susquehanna and northern Bay waters

move southward, they become less

turbid. The temperature also changes

because increasing time is available

for mixing, sensible heat exchange and solar insolation. From the Patuxent

River to the Rappahannock, a pool of clear and cooler water persists. This

region is relatively deep and furthest

from sources of suspended sediments

and freshwater; hence it has low turbidity and the uniform and cooler

temperatures. Plumes emanating from

the Rappahannock and York Rivers are advected southward along the western

shore. This water appears to merge

with the James River plume and the

282
combined unit moves seaward along the coast around Cape Henry. Most of the exchange of water with the

shelf probably involves the central

and eastern portions of the lower Bay. This Bay water closely resembles the

shelf water in turbidity and

temperature, hence the exchange cannot

be readily seen in the images. Our study of Chesapeake Bay and the adjacent shelf illustrated the utility

of the approach in a complex

environmental system. Adaptation to other areas is similar.

13 . CONCLUSIONS

Satellites can provide important

environmental information on Chesapeake Bay and the adjacent shelf. Using the satellite data, we can

obtain a synoptic view of sea surface

temperatures, particularly during the

day, calculate the reflectances caused

by suspended solids, and determine the presence and strength of phytoplankton

blooms. Some characteristics of surface circulation can be inferred

from the distributions of temperature

and reflectance. With the presence of well defined features, the satellite

data can be used to determine actual

circulation, as shown by the movement

of Eddy 12 on the shelf.

At present, satellites can provide

accurate temperatures within 1 °C. Reflectance values give an indication

of the relative concentration of suspended solids; however, estiamtes

of atual concentrations will depend on our ability to calibrate in situ data

against remote measurements of reflectance. Similarly, the pigment

index requires additional calibration

in order to make it more widely

applicable. However, the use of reflectance and a pigment index show

the potential ability of satellites in studying and monitoring estuaries.

The temperature data can be used to initialize the surface temperature

component of the MECCA circulation

model. In addition, it can be used to verify the temperature distributions

predicted by the model (section

6.6.3). Similarly, when material

transport is added to the model, bloom

and reflectance distributions could be

compared with the model predictions.

The synoptic coverage and availability

of satellite data fill a void caused

by the logistics and expense of conventional oceanographic

measurements

.

We have shown that satellite data together with other physical

parameters (meteorlogic and oceanic)

are useful indicators of fisheries.

By understanding the major oceanic

features (cold and warm pools, fronts, and productive regions) , and relating

them to fishery information, managers

and policy makers will have additional

information to based their decisions.

14. REFERENCES

1.

Gagliardini, D.A., H. Karszenbaum,

R. Legeckis, and V. Klemas. 1984. Application of Landsat MSS, NOAA/TIROS

AVHRR, and Nimbus CZCS to study the la Plata River and its interaction with

the ocean. Remote Sensing of Environment 15:21-36.

2.

Gordon H.R., D.K. Clark, J.W. Brown, O.B. Brown, R.H. Evans, and

W.W. Broenkow. 1983. Phytoplankton

pigment concentrations in the Middle

Atlantic Bight: comparison of ship

determinations and CZCS estimates.

Applied Optics 22: 20-36.

3. Munday, J.C. and T.T. Alfoldi.

1979. Landsat test of diffuse reflectance models for aquatic

suspended solids measurements.

Remote Sensing of Environment

8:169-183.

4. Strong, A.E. and E.P. McClain.

1984. Improved ocean surface temperatures from space—comparisons

with drifting buoys. Bulletin

American Meteorological Society

62(2) : 138-142.

5. Stumpf, R.P. and M.A. Tyler.

Satellite detection of bloom and pigment distributions in estuaries. (Submitted to Limnology and Oceanography ^

.

6. U.S. Department of Commerce. 1982. Tidal Current Tables, Atlantic Coast

of North America. NOAA, National

Ocean Survey, Rockville, MD.

283
Table 1. Spectral bands measured by AVHRR, CZCS, and Landsat MSS. Numbers denote sensor channels.

m . 4

1

visible | .5 .6 .7

1 1 1

Spectral Bands near-IR | .8 .9 1.0

1 I 1

thermal-IR

3.5 11 12 13

1 III

AVHRR 1 2 3 4 5 CZCS 1 2 3 4 5 6 Landsat MSS 4151 _6_J 7

Table 2. Characteristics of Satellite Sensors. Spatial Maximum Image Coverage

Resolution Scan Angle Width Period

(km) from vertical (km) fdavs) AVHRR 1.1 56 2000 1/2 CZCS 0.82 38 1500 6 Landsat MSS* .08 6 180 18**

* Landsat: Thematic Mapper, which has a spatial resolution of 30

m, began operating in fall of 1982 . ** In 1979, the period was1 9 days. Starting with Landsat 4 in fall 1982, the period between images changed to 16 days.

Table 3. Imagery of Chesapeake Bay used in this ;study.

Date Time CEST) Satellite Orbit Number

10 April 82 0800 NOAA-6 14477

1148 CZCS 17475

12 April 82 0805 NOAA-7 04143

1045 CZCS 17502 1429 NOAA-7 04141

14 April 82 0239 NOAA-7 04162 0810 NOAA-6 14534

1410 NOAA-7 04169

15 April 82 0230 NOAA-7 04176 0740 NOAA-6 14548

1137 CZCS 17544

1400 NOAA-7 04183

16 April 82 1157 CZCS 17558

1345 NOAA-7 04159

18 April 82 0810 NOAA-6 14591

1050 CZCS 17585

23 April 82 0755 NOAA-6 14662 1400 NOAA-7 04296

24 April 82 1058 CZCS 17668

284
DISCRIMINATION OF FLOODED AREAS CAUSED BY TSUNAMI OR

STORM SURGE USING SATELLITE DATA

by Atsushi Takeda^ Yoshinobu Tsuji^ Hiroshi Takahashi^

ABSTRACT

Sea water will damage vegetation in flooded areas

caused by a tsunami or a storm surge. An examination to discriminate the flooded areas macroscopically of vegetation changes appearing in satellite images was made from the Tsunami

caused by the Nihonkai-Chubu Earthquake in 1983.

1. INTRODUCTION

A macroscopical method using satellite data is used to detect flooded areas caused by tsunamis

or storm surges. Traces of sea water from the flood remain for several months; vegetation

changes from washing away, burying, and withering. These vegetation changes appear as spectral images of an earth surface observed

from a satellite. Most parts of the earth surface are observed periodically by high- resolution spectral image sensors such as MSS

and TM on LANDSAT satellites. A careful comparison of two images observing the same land area

before and after a disaster provide useful information about traces of remaining sea water.

This report addresses the flooded areas from the tsunami caused by the Nihonkai-Chubu Earthquake

on May 26, 1983. The tsunami struck the coast

of Minehama Village, Akita Prefecture in Northern Japan. Data used in this investigation

was provided by MSS image data of LANDSAT.

2. DATA

Two scenes of satellite data were used for analysis of the examination. One is Scene

No. 108-32 of LANDSAT-4 observed on July 12, 1983, one month and a half after the tsunami.

The other is Scene No. 116-32 of LANDSAT- 3

,

July 8, 1982, one year earlier. It was

selected to compare the area, assuming that seasonal conditions of vegetation was almost equivalent except for the stricken areas. Each

scene covers the areas required for analysis.

Data of each scene consist of 4 spectral images; Band 4 through Band 7 , respectively. In the analysis, images from Band 5 and Band 7 were

used. These two bands are important because

they reveal vegetation changes. Band 5 (0.6-0.7^m) involves an absorption band and Band 7 (0.8-l.l^m) a reflection band of chlorophyl in green plants. The other bands are

less important. The ground resolution of each

image is 80m. The ground data on actual flooded areas were

referred to those reports of the field surveys

made immediately after the tsunami by several

groups. Fig. 1 shows a map of flooded areas based on the data.

3. ANALYSIS

To perform an objective description of differences between two images of each band obtained

before and after the tsunami, quantitative

analyses were made. Pre-processings, geometric

corrections and optical sensitivity adjustments

were performed, so the two images of each band coincident geometrically and optically except

for the stricken areas. A sampling area P for an analysis, of 10km x 25km was cut out from an original image of 180km x 180km.

As the result of pre-processings, two sampling

of band i (i=5, 7) are equivalent, if there is no change over the surface of sampling area, and pixels (pixel size; 80m x 80m) in each sampling

image corresponding to a ground point p(x,y) should have equal brightness. Consequently,

the difference between the^Bj(x,y) should be zero. On the other hand, if^Bi(x,y) is not zero, it is suggested that there should be some change at the ground point corresponding to the pixels.

Thus, a subtracted image between two images of each band before and after the tsunami are shown

in Fig. 2, where each pixel has a value of^B^.

Here . /\ B-> is defined as the difference of a value before the disaster from a value after it. For a more quantitative expression, a standard deviation of,^Bi, <7"b is calculated over a whole sampling image, and the pixels which lie on the land and are selected as "significant difference pixel" = SDP. Fig. 3 a and b are a distribution map of SDP, where a SDP of and a SDP of 1 A 3^172 0^ are indicated in a andO, respectively.

4. DISCUSSIONS

The false colored image after the tsunami shows

3 bright spots along the coast of Minehama

Village. These spots are not clear in the image before the tsunami. On the other hand, it is found that positions of these spots coincide exactly with those of three large flood areas in Fig. 1.

^University of Tokyo, Hiratsuka Branch, Japan ^National Research Center for Disaster Prevention, Sakura, Ibaraki 305 Japan

285
The subtracted images of Band 5 and Band 7 have bright and dark spots, respectively, at the positions the false colored image has spots; a bright spot is positive, and a dark spot is negativei^B^

.

Fig. 3 illustrates these facts clearly. The figure also shows that distributions of SDP are concentrated in a zone along the coast line; it does not expand into inland areas. Almost all the SDPs in the two figures of Fig. 3 lie within the flooded areas in Fig. 1. Furthermore, the superimposed map of the two figures shows that distribution patterns of SDP are very similar to those of actual flooded

areas except very narrow zones close to the shoreline

.

Band 5 is an absorption band, and Band 7 is a reflection band of plants. When plants wither, absorption and reflection decrease; the brightness of Band 5 increases and that of Band 7 decreases. The results of the analyses

agree to this inference.

A problem still remains. Few SDPs appear in a zone close to the shoreline where the tsunami

intruded, and many black-pine trees died. Perhaps some plants on a dune are durable against

sea water, wither several months after the flood, or revive soon after the damage.

5. CONCLUSION

Satellite data are useful for discriminating

traces of flooded areas caused by a tsunami or a storm surge. This method will be used for field surveys immediately after the disaster. More similar analyses based on several satellite observations at different times are needed. The satellite data must keep pace with the vegetation cycles for more exact discrimination.

Only MSS images were available when the Nihonkai-Chubu earthquake occurred. Today, higher resolution pictures of Thematic Mapper

image are available permitting greater detailed

analyses of flooded areas.

6. ACKNOWLEGEMENTS

Satellite data used here were provided by courtesy of the National Space Development

Agency. Thanks also to the local governments of Akita Prefecture and Minehama Village for their cooperation.

7. REFERENCES

(1) Akita Local Government (1984); Documents of

the Nihonkai-Chubu Earthquake in 1983, Akita

Local Govt. 420 pp. in Japanese.

(2) Forestry Div. of Akita Prefecture (1984); Documents of the Nihonkai-Chubu Earthquake,

Green Waterbreak, 20 pp. in Japanese.

286
(NlHONKAl)

287

of

Nihonkai-Chubu

Earthquake

along

Subtracted

images

(a)Band

the

coast

of

Minehama

Village,

Akita

(b)Band

7

Prefecture

.
Fig. 3. Distribution map of SDP, (a) Band 5, (b) Band 7

288
FAULT DISPLACEMENT SEICHE WAVES ON

INLAND RESERVOIR AND LAKES

By Robert L. Deweyl

and

Karl M. Disel

ABSTRACT

Vertical displacements on faults that

have been identified along the bottom

and in the vicinity of inland lakes and reservoirs could induce seiche waves on the water surface that would create

damage along the shoreline. Prediction

of fault displacement seiche wave

heights is important at the Bureau in order to assess the safety of dams exposed to this hazard. This paper pre- sents the Bureau's experience with com- puter analyses to predict seiche wave

heights at a few dams potentially

impacted by fault displacement waves.

Wave characteristics and the parameters

that influence the ultimate wave height

at the dam are discussed. Evaluation of the wave's impact at the dam is also

presented. Areas of analytical uncer- tainty are identified with recommendations for an appropriate conservative

approach. Further research is suggested

to reduce these uncertainties and advance the state-of-the-art.

KEYWORDS: Fault Displacement; Seiches;

Sei smi c Risk; Waves

1 . INTRODUCTION

If vertical displacements are expected

on a fault located along the bottom or

in the vicinity of a lake or reservoir,

engineering designs at the water's edge

need to account for the potential for extremely large waves impacting these structures during an earthquake. These waves can be called seiche waves.

Seiche waves are created the moment

vertical offset on the fault takes place.

In an attempt to reach equilibrium, the volume of water that becomes higher

immediately after the displacement prop- agates toward the lower end of the reservoir. Characteristics of the wave

are primarily a function of the magnitude of the fault displacement, the location of the faults, the shape of the reservoir, the bathymetry of the reservoir bottom, and the reflectance of the

shoreline, among other things. An appropriate model of these factors along

with a good mathematical solution can lead one to predict the wave expected to impact the shoreline. Ever since this type of event resulted in overtopping

Hebgen Lake Dam, Montana, the U.S. Bureau of Reclamation has attempted to predict fault displacement seiche wave heights at projects where that potential exists.

2. METHODS OF ANALYSIS

The computer program LSWAVF [1] was used

to assist in the prediction of fault displacement seiche wave heights that may occur. The program was developed to simulate the propagation of waves throughout a reservoir or lake using numerical techniques. The waves that

can be analyzed may be generated by a landslide entering the basin or by a bottom fault displacement. The equations of motion used in LSWAVE are the nonlinear, depth-averaged continuity

and momentum partial differential

equations derived from Na vi er-Stokes equations. Calculations are made for a variable-depth basin using an explicit

finite difference method (a leapfrog

scheme) for solving the partial dif- ferential equations. The forcing mechanism of the wave is incorporated into

the continuity equation. A second forcing mechanism, drag induced water motion, is incorporated into the momentum equations. The equations and their solutions used in LSWAVE are presented

in reference [1]. Varying boundary con- ditions are permitted to treat dam over- topping and shore reflectivity.

A body of water is modeled in LSWAVE by

assigning water depths to the nodal points of a uniformly spaced grid. Land within grid boundaries is represented by

1 Civil Engineer, U.S. Bureau of Reclamation, PO Box 25007, D-230, Denver, Colorado 80225.

289
by grid points that are assigned depths

of 0.0. Fault movement is described by two lines: one of maximum displacement

(the fault trace) and one of zero displacement (the hinge line). The reservoir floor instantaneously drops at time zero and displacements are calculated by linear interpolation at grid

points located between the fault trace

and the hinge line. The water surface

grid points are displaced by an identical amount creating a discontinuity in the vicinity of the fault trace. Shoaling and reflection are accounted

for by a Manning roughness coefficient

and a CED Coefficient of energy

d i s si pat i on ) , Unless waves are totally reflected by the shoreline, wave energy is dissipated

to some degree when it reaches the edge

of the reservoir. One CED is selected

to describe the dissipative capability

of the entire reservoir shoreline. The

CED can vary from 0.0 for totally

reflective boundaries to 1.0 for totally

transmissive boundaries. The slope and the roughness of the shoreline contribute to the CED. The sensitivity of the results to varying CED's should be studied for each analysis.

A grid of squares composed of up to

8,000

nodal points is required by LSWAVE

to describe the reservoir bathymetry.

There must be enough grid points to blanket the area of the reservoir and spaced such that at least six points are used to describe the expected wave- length. Thus, a square grid of a uni- form spacing is applied to reservoir and shoreline features of all shapes and sizes. Seismic movements modeled by the program

LSWAVE could be considered static type rather than dynamic. The program

assumes an instantaneous rupture along a fault on the lake bottom and calculates

the resulting movement of the reservoir

water surface for a given amount of time afterward. The instantaneous displacement of the water surface is an approximation of the effect of the dynamic

release of energy over the entire lake bottom during the earthquake process.

The assumption of instantaneous rupture

could be slightly conservative because

these displacements may take several seconds to propagate across the lake. Additional reverberant motions over a significant area surrounding the fault

are not considered in this program.

What is offered by the program is an approximation of the earthquake source.

Other input r eq u i r emen t s , such as Manning's roughness coefficient, length

of time history and time step, output

options, etc., do not significantly

influence the results and will not be discussed here. The following section presents cases that the Bureau has studied which demonstrate the use of LSWAVE for the fault displacement seiche

wave analysi s .

3. CASE HISTORIES

3 . 1 Lake Tahoe Dam (California/Nevada)

3.1.1 Description of Problem

Lake Tahoe occupies a graben that has been downdropped during Quaternary time relative to the Sierra Nevada to the west and the Carson Range to the east

(see Fig. 1 and table 1)[2]. If a fault displacement were to occur, the lake bottom would be expected to drop like a trapdoor, hinged along another parallel

fault on the opposite side of the lake. Keeping with theories of mass conservation, the lake water surface will simultaneously drop by the same amount

to one side of the ruptured fault. Waves will then travel from both directions perpendicular to the ruptured

fault to fill in where the water surface

has dropped. Shortly after this hole begins to fill, waves will move out toward the shore and the dam. The distance that the wave travels, lake bottom topography, and shore outline

contribute to the form of the wave.

Lake Tahoe Dam is located on the north- western shore of the lake basin where

the lake bottom is relatively shallow

and slopes are relatively gentle toward

the middle of the lake. These factors

can cause a wave to increase in height

as it moves toward the dam. In fact, these effects, referred to as "shoaling," may produce a wave at the dam greater than the original fault

d i spl acement

.

There are three possible cases of fault displacements within the shores of Lake Tahoe. The first is where the lake bottom just east of a fault running along

the western side of the basin drops

about 7 feet and decreases linearly eastward to 0 foot at a hinge on a fault running along the eastern side. This case is called "WE," the first letter standing for where the displacement

occurs, the second letter standing for the hinge. The second case is just the opposite, "EW," where the fault along

290
the eastern side is displaced about

7 feet, dropping the lake bottom just

west of the fault by that amount and decreasing linearly to 0 foot at the hinge located on the fault running along

the western side. The third case

involve dropping of the lake floor

about 5 feet just south of a fault run- ning along the north side. The dis- placement decreases linearly to 0 foot

at the south edge of the lake basin on a hypothetical hinge formed at the bottom

grid line. This case is called "NS." Since Lake Tahoe is about 72,000 feet by

115,000 feet, the grid pattern selected

was 69 by 110 spaces. Each grid point

was 1,050 feet apart so that about six points would be used to model the expected wavelength of 6,200 feet. No more grid points or closer spacing was possible within the limits of LSWAVE's

present computational capability.

Various CEO's were selected to describe

the percentage of the wave's energy

absorbed by the shoreline. Selection of

an appropriately conservative CEO is discussed later.

3.1.2. Results and Interpretations

The results of the various computer runs are presented in table 2. Maximum wave heights were evaluated for all possible

seismic cases and many values of CEO. The prediction of the heights of waves expected at any point along the shoreline is dependent on the interpretation

of the maxi urn wave heights recorded at numerous other grid points in the vicinity of the point of interest.

Various interpretations can be made to predict the wave height at the dam depending on which grid points are selected and how they are examined.

Fig. 2 shows numerous grid points along

the shore of Lake Tahoe used to model

the vicinity of the dam. Points on land

are identified as X, points within the

lake are identified as 0. The dam is located near lake points labeled G and E . Many interpretations were studied, but only the four shown in table 2 were con- sidered reasonable. Interpretations

that included corner points G, P, or U are not included in table 2 because the results at these points were anomalously

high. Either numerical problem' s or an artifical compounding occurred at these

points. Since the shoreline does not actually have sharp corners like this,

averaging surrounding points or choosing

points not at a corner would afford a better interpretation. During an inde- pendent review, the U.S. Corps of Engineers [3] offered two of the inter- pretations shown in table 2. Relative

to the Bureau's interpretations, the

Corps' interpretations lead to higher

expected waves for seismic cases WE and

EW, and lower expected waves for seismic

case NS. This is because the maximum

wave heights increase from south to north and the Corps of Engineers' inter- pretations include more points that are north of the points used in the Bureau's interpretations.

Selecting just one interpretation over

the others to best demonstrate the waves expected to impact the dam is difficult.

Instead, it is prudent to formulate conclusions based on giving equal cre- dence to the four reasonable

i nterpretati ons . The results in table 2 are fairly sen- sitive to the CED. The amount of wave energy reflected by the shoreline is dependent on:

• The slope of the shoreline.

• The roughness of the shoreline.

• The steepness of the incoming wave

(hei ght/1 ength) . Fault displacement seiche waves are extremely long such that steepness is always very low. A relationship given

in the Shore Protection Manual by the

U.S. Corps of Engineers [4] between the wave steepness and the slope of the shoreline finds that at such small wave steepnesses, the effect of steepness

becomes negligible with respect to the effect of the slope of the shoreline on energy dissipation. Although the average angle of Lake Tahoe's entire

shoreline was calculated to be about

2.25 degrees, the angle of the northern two-thirds of the east shore and the southern one-third of the west shore is quite a bit higher (locally up to

10 degrees). Nevertheless, when accounting for roughness, a conservative

estimate of the CED was determined to

be 0.4. 3.1.3 Conclusions at Lake Tahoe

The highest value of maximum wave height selected from the four reasonable interpretations at a CED value of 0.4 offers a conservative estimate of the fault displacement seiche wave expected

to impact Lake Tahoe Dam during a

291
seismic event. The controlling seismic

event is a Richter magnitude 7.0 earthquake, displacing about 7 feet along the eastern fault and hinging along the western fault. The reasonably conser- vative upper bound value of the maximum

fault displacement seiche wave height

expected to impact Lake Tahoe Dam during

that event is 11 feet. The buildup of such a large wave can be attributed to a considerable amount of shoaling as the wave approaches the dam. Since the dam is comprised of both a concrete gate structure and two embankment fills that tie the concrete structure into the riverbank abutments, there

are many potentially detrimental effects

that such a wave can have on the structure. The two most critical effects

were found to be sliding instability

caused by the horizontal hydrostatic

pressures of the wave tending to push

the concrete structure downstream and washing away of the embankments by over- topping. The undrained behavior of the foundation soil was found to be dilative. Therefore, in the event of a sud- den load caused by this fault displacement wave, the horizontal shear strength

of the soil under the concrete structure

will increase as the soil begins to strain so to resist the increased

hydrostatic pressures. Surface protection of the embankments has been designed to prevent erosion during the temporary overtopping. Once these improvements have been constructed. Lake Tahoe Dam will be able to survive the expected seiche wave created by fault displacement through the water.

3.2. Jackson Lake Dam (Wyoming)

3.2.1. Description of Problem

Jackson Lake Dam is a composite earthfill and concrete dam constructed

between 1910 and 1916 on the Snake River

in western Wyoming. The dam consists of

a concrete gravity section with a miscellaneous fill embankment to the south and a hydraulic fill embankment to the north. The program LSWAVE was used

to estimate the magnitude of fault displacement generated waves and the effect these waves have when they reach

the dam. The study was conducted to determine if the dam's crest elevation

should be raised to accommodate free- board loss from these waves.

Most recent fault trace locations are shown in Fig. 3. Trace 1 has been observed along the shoreline and

1 mile to the west of Jackson Lake and

it is considered to be the most likely

location for future rupture. Traces 2 and 3 were inferred from Jackson Lake bathymetry but no evidence of these

faults was found outside the reservoir.

There is inconclusive geophysical evi- dence for the existence of the Hermitage

Point fault trace also shown in Fig. 3. Detailed study of large-scale aerial photographs along the postulated trace suggests that surface displacements have

not occurred in the last 20,000 years.

However, movement along this trace was thought to have the potential for creating the worst-case situation for fault displacement generation of waves,

so it was included in this study.

Relationships between maximum fault displacements, net geodetic change, and distance from a fault trace are pre- sented in Fig. 4 [5]. These rela- tionships were developed based on displacement data associated with historic normal faulting in the Western

United States along with observation of the basin width and the configuration of the Huckleberry Ridge Tuff in the Jackson Lake area. The maximum credible

surface displacement along the Teton

fault (traces 1, 2, or 3 on Fig. 3) was estimated to be 22 feet with the east

side moving down relative to the west

side. The maximum credible surface displacement along the Hermitage Point fault was estimated to be 5 feet, again down to the east. The net elevation

change at Jackson Lake Dam and the loca- tion of the hinge lines for each of these events can be determined from

Fig. 4. For the Teton fault, the net elevation change at the damsite would be down 13 feet and the hinge line would be located approximately 5 miles to the east of the dam. For the Hermitage

Point fault, the net elevation change at the dam would be down 2 feet and the hinge line location would be approximately 1 mile to the east of the dam. Jackson Lake was modeled using a grid

67 divisions wide by 99 divisions long. The grid spacing was approximately

668 feet square and seven grid points

were used to represent the dam. Before

the program could be run successfully,

the reservoir boundary required

smoothing and islands had to be removed

by assigning 1-foot depths to grid

points located within the island boun- daries. The time step used was

4 seconds and the CEO was varied from 0.30 to 0.75.

Best estimates for locations of the Teton and Hermitage Point faults were chosen for initial modeling attempts.

Sensitivity to the choice of fault and

292
hinge line locations was then tested by

arbitrarily moving these lines parallel,

2,000 feet in either direction.

3.2.2. Results and I n t e r p r e t a t i o n

s

In each fault movement simulation, the estimated reservoir behavior was a com- bination of effects generated by a discontinuity in the water surface at the fault trace and by the tilting of the reservoir basin away from the dam. Initially, the water surface near the

dam fell as water moved away from the embankment due to reservoir tilting (see

Fig. 5a). At some time later, depending

on the distance of the fault trace from

the dam, the fault displacement induced wave reached the vicinity of the dam and the water surface rose again. The water

surface then oscillated until all energy

was dissipated at the shorelines or until the simulation was stopped (see

Fig. 5b).

Varying the CEO from 0.30 to 0.75 had little effect on results at the dam (see

Fig. 6). Therefore, a value of 0.60 was used throughout the analysis.

Figs. 7 and 8 are plots of the time history of displacement for the water

surface at a grid point next to the dam. Results of fault displacement along the Teton fault are presented in Fig. 7, while Hermitage Point fault displacement

results are shown on Fig. 8. Water sur- face displacements represent the change

in elevation from the assumed original water surface elevation of 6769 feet.

The f a ul t- i nduced elevation change of the dam crest is equal to the initial water surface elevation change at time

zero. Therefore, the net difference

between displacement at any time and displacement at time zero is the change

in freeboard. The initial freeboard was assumed to be 11 feet. Figs. 7 and 8 show the effects of changing the locations of the hinge

lines and fault traces. Changing the location of the hinge lines had little effect on water surface displacement at the dam. Changing the location of the Teton fault significantly affected water displacement but changing the location

of the Hermitage Point fault did not. For the best estimate location of the Teton fault, the minimum freeboard was

6.5 feet. When the fault was arbitrarily located 2,000 feet to the east, the minimum freeboard was

5.5 feet. For the best estimate loca- tion of the Hermitage Point fault, the minimum freeboard was 6.1 feet and changing the location 2,000 feet to the

east only decreased the freeboard by 0.5 f oot

.

In order to determine if boundary con- ditions at the grid points which repre- sented the dam caused computational

errors, water surface displacement versus time data were obtained for all grid

point locations in the vicinity of the

dam. Only slight differences were noted

in the results. The shapes of the displacement versus time curves differed

from grid point to grid point with some peaks sharper than others, but the periods were virtually identical and the maximum amplitudes were within approximately 0.5 foot. 3.2.3. Conclusions at Jackson Lake Analytic results from studies presented

in this report indicate that fault displacement induced waves will not overtop Jackson Lake Dam if the dam is rebuilt to its current crest elevation

of 6780 feet. However, there are uncer- tainties in the results due to limita- tions in the ability to predict the

actual fault displacement mechanism, in the ability of the computer code to ade- quately model these mechanisms, and in the small scale wave behavior in the vicinity of the dam. Constructive

interference from several waves possibly

generated by a multiple fault displacement event could conceivably create a wave that would overtop the dam but it is believed that the duration of the overtopping would be significantly

shorter than the 18-minute period waves predicted in this report and that the overtopping would not be a recurrent phenomenon. The uncertainty of a con- ceivable dam failure caused by a fault- displacement induced wave overtopping

the dam can be mitigated by placing

slope protection material on the downst ream slope.

4. CONCLUSIONS

Successful, conservative predictions of fault-displacement seiche wave heights

can be made using the computer program

LSWAVE. The amount of conservatism may tend to be high due to numerous uncer- tainties inherent in the analysis. Additional research and computer pro- gramming can eliminate some of these uncertainties and render more accurate predictions. Improvements to the program LSWAVE which are technically

feasible at this time include:

• Allow for different types of grids:

293
Va ri abl e recti 1 i near

C u rv i 1 i nea r Patched or subgrid features

" Develop a new solution scheme with dynamic time stepping.

• Allow for variable CED's along

different sections of the reser» voir shorel i ne.

• Allow for a more dynamic type of fault displacement and reservoir

bottom drop. These improvements were beyond the scope

of the analyses presented in this paper. Without any of these improvements, con- servative choices were made where the uncertainties exist. Areas within the LSWAVE model where conservatism may be introduced included:

• Selection of lower values of the

CEO.

• Selection of the most critical

sei smic case.

' Selection of the most critical interpretation of maximum wave heights at certain strategic grid

points.

• Selection of the maximum fault offset possible and assume it occurs suddenly along the entire

fault line resulting in a sudden drop of the reservoir bottom. Accurate estimations of these parameters

along with improvements to the program

will lead to better, more reasonable engineering designs along a shoreline

susceptible to fault displacement seiche

waves. Nevertheless, the Bureau has demonstrated that fault displacement

seiche waves can be predicted so that

safe designs can be built accounting for what otherwise would be a catastrophic

event

.

[2] U.S. Bureau of Reclamation,

Memorandum, Subject: "Preliminary

Se i smot ec

t

0 n i c Conclusions for Lake Tahoe Dam - Newlands Project,

Cal i forni a-Ne vada February 10, 198A„

[3] U.S. Corps of Engineers, "Fault Displacement Seiche Wave Analysis

Review," Coastal Engineering Research

Center, Memorandums for Record, 26 April 1985 and 6 August 1985.

[4] U.S. Corps of Engineers, "Shore Protection Manual," U.S. Army Coastal Engineering Research Center, 1977 and 1984.

[5] U.S. Bureau of Reclamation,

"Seismotectonic Study, Jackson Lake Dam, Minidoka Project, Wyoming,"

Seismotectonic Report 83-8, Engineering

and Research Center, Denver, Colorado,

September 1983.

5. REFERENCES

[1] Chiang, W., D. Divoky, P. Parnicky,

and W. Wier, "Numerical Model of Landslide-Generated Waves," Tetra-Tech

Report Number T-3427, Pasadena,

California, November 1981. (Note:

Program LSWAVE modified by Julie Heim,

May 1984).

294
Table 1. - Faults within Lake Tahoe [2]

Estimated Am ount 0 f Sense of MCE rupture length maximum surface

Fault displ acement (Ms) ( km

)

displacement ( m)

Tahoe (west side) No rmal , down to east 7 40 (24.9 mi

)

1.8 - 2.1 (6.9 ft) Tahoe (east side) No rmal , down to west 7 40 (24.9 mi

)

1.8 - 2.1 (6.9 ft) Tahoe ( north side) No rmal , d own to 6-3/4 20 (12.4 mi

)

1.2 - 1.5 (4.9 ft) southeast

Table 2. - Summary of maximum fault -displ acement seiche wave heights

expected at Lake Tahoe Dam for di fferent interpretations and various sei smic cases,

Bureau of Reclamation ' s Co rps of Engineers' Seismic cases interpretation interpretation [3] 1,050 ft east Th ree points

Fault Hinge CED Dam of dam Dam surrounding dam

(E) (F) (H) (E, H, and I)

W E 0.0 7.8 7.8 13.0 11.3

W E 0.2 6.6 6.2 9.2 8.6

W E 0.4 6.9 4.2 9.8 8.4

W E 0.6 6.0 3.1 5.7 5.9

W E 0.8 4.1 2.3 4.3 4.1

E W 0.0 7.5 11.0 15.0 15.2

E W 0.2 6.7 8.7 9.9 10.8

E w 0.4 5.6 7.7 11.0 9.2

E w 0.6 4.7 7.1 6.1 5.9

E w 0. 8 3.7 7.1 7.1 5.6

N s 0.0

N s 0.2 4.9 5.5 5.5 5.1

N s 0.4 4.7 3.9 2.6 3.3

N s 0.6 2.6 1.9 2.2 2.0

N s 0.8 1.7 0.80 0.90 1.2

295
Flgur« 1. Bathymetric map of Lake Tahoe showing location

of Lake Tahoe 0am and inferred Tahoe fault

locations.

\ 296
Figur* 2. Grid points aiong tho shorsiino noar Laka Tahoa Dam

297
298
FItur* f. Wat«r surfac* dlsplacamant vs. tlm#

for various cosfficiants of ansrgy dissipation.

301
o

(133J Nl)lN3^N30V1dSia

(yi Q

o

o

Ld00

FIflur* 7. Wat«r turfaea dlsplacamant vs. tiaia for various locations of tha Harmitaga Point fault traca and hinga lina.

302
Fiflur* 8. Wat«r surfaca 8ia^laaaaiant vs. tint# for various locations of tha Taton fault traca and hinga Una.

303

THEME IV

U.S.- Japan Cooperative Research Program

U.S.-JAPAN COORDINATED EARTHQUAKE RESEARCH PROGRAM ON REINFORCED MASONRY BUILDINGS -SEISMIC CAPACITY OF REINFORCED MASONRY WALLS AND BEAMSBY Shin Okamoto*, Yutaka Yamazaki*, Takashi Kaminosono*

Masaomi Teshigawara*, and Hisahiro Hiraishi*

ABSTRACT

As part of the U.S. -Japan cooperative research

program on masoniry building structures, work

performed included testing many reinforced

masonry walls and beams to investigate

experimentally how the amount of shear rein- forcement, axial stress, shear span ratio, reinforced concrete (R/C) slab and transverse

walls affect shear and flexural behaviors of masonry walls and beams. To resist severe

earthquakes, strength and deformation capacity

are the most important factors for medium- rise masonry building structures. The major

findings follow:

1) , Increase in amount of shear reinforcement

and spiral reinforcement was effective on deformation capacity.

2) Maximum shear stress increased with

increasing of the axial stress.

3) Under the shear failure, the deformation

and strength capacity of the walls were

improved with the existence of the transverse walls.

4) The shear and flexural strength can be estimated by the empirical formulas specified for R/C.

1. INTRODUCTION

During 1984 and 1985 the Building Research

Institute, Ministry of Construction, conducted

seismic behavior test of masonry walls and

beams. These tests were carried out under the

U.S. -Japan Coordinated Earthquake Research

Program on Masonry Building Structures. The purpose of this study was to investigate

experimentally how the amount of shear rein- forcement, axial stress, splices, confining of compression zone, transverse wall, shear span

ratio, and R/C slab affect shear and flexural behaviors of masonry walls and beams. Thirtyfive specimens were used in this study. They

included 25 masonry blocks, 7 clay bricks, and

3 reinforced concrete specimens.

*Building Research Institute Ministry of Construction

2. SPECIMEN AND TEST SETUP

2 . 1 Materials

The concrete block masonry unit used for the specimens consists of two webs and two open

ends with nominal dimensions of 20 cm thick, 40 cm wide, and 20 cm high. The clay brick

masonry unit has the same shape as the concrete block unit, but the nominal dimensions are 20 cm thick, 30 cm wide, aijd 10 cm high. Those masonry units are shown in Fig. 1. The material properties of the specimens and the companion prism strength are tabulated in Table

1 .

2 . 2 Specimens

Typical specimens are shown in Fig. 2. The specimens were constructed in running bond with

a 50% overlap unit in alternate courses, and

are fully grouted by concrete. The details of specimens are listed in Table 2. The specimens

WTTl, WTT2 and WTCl have transverse walls. The beam specimens GFl, GF2, and GSl have R/C slab, they are 15 cm thick and 1 meter wide on each

side of the beam.

2 . 3 Test Setup

The test setup (Fig. 3) employed in this study

was the Building Research Institute's (BRI) Two-Directional Test Facility (BRI/TTF) . The BRI/TTF is a computer-controlled loading

apparatus which can apply in-plane force and displacement in three degrees of freedom at the top end of a specimen, there are two translations and one rotation. The bottom end of a specimen is fixed to the test floor. Specimen

loading is accomplished by three servo- controlled hydraulic actuators which receive

their instructions by computer-generated

commands. In this test, two vertical actuators

were controlled by the computer to restrain the rotation of the top end of the specimen and to keep constant vertical compressive stress.

2.4 Loading Rule

In principle, specified loading history was applied to each specimen. It consists of 1) one-cycle to the displacement corresponding to the elastic limit, 2) two-cycles to a drift

angle of 1/400, 3) two-cycles to 1/200, 4) two- cycles to 1/100, and 5) one -cycle to 1/50.

307
3. TEST RESULTS OF WALLS

Figure 4 shows crack patterns at the final

stage of some specimens. Table 3 lists the mean shear stress and drift angle at the maximum strength. This table also lists the ultimate stresses calculated by equations (1) and (2) described in the same table.

3.1 Effects of Axial Stress (WFl . WS4. WSNl

.

WSN2) Figure 5 illustrates the relationships between

the shear stress and the axial stress. The maximum shear stress increases from 16 kg/cm^

to 23 kg/cm^ with increases of the axial

stress

.

The specimen WFl with an axial stress of 5 kg/cm^ showed, 1) flexural yielding occurred

about at drift angle of 1/500, 2) maximum

strength occurred about at drift angle of

1/250, and 3) a 20% decreasing of the maximum

shear stress was observed about at drift angle

of 1/100. The specimens with the axial stress

of 20 kg/cm^ and more showed, 1) maximum

strength occurred about at drift angle of

1/500, and 2) a 20% decreasing of the maximum

shear stress was observed about at drift angle

of 1/250.

3.2 Effects of Shear Span Ratio (WSl. WSRl.

WS4. WSR4. WS7. WSR7

Processes of crack propagation of each specimen had a similar tendency to a drift angle of

1/800. At first, flexural cracks were observed

at the top and bottom parts of the specimens. Then shear cracks occurred at the wall corners

and those were developed gradually with

increasing number of loading cycles. When the drift angle reached to 1/400 (1/800 for the specimen WSl) , the diagonal shear cracks

occurred. Figures 6 and 7 show the effect of shear span

ratio (M/QD) on the shear stress vs. drift

angle relationships. The maximum shear

stresses of these specimens stand in the order

of WSR1>WS1, WSR4>WS4, and WSR7-WS7. The specimens WSl and WSRl showed approximately 20% higher in shear stress compared with the specimens WS4 and WSR4, and also 30% higher

than the specimens WS7 and WSR7 . The specimens

WS7 and WSR7 which have a shear span ratio of

1.14, failed in flexure and showed good deformation capacity. The concrete block

masonry wall WS7 has a larger deformation

capacity than clay brick masonry wall WSR7 . In

the case of the specimens which shear span

ratios are 0.45 and 0.76, the clay brick

masonry walls (WSRl and WSR4) were superior to the concrete block masonry walls (WSl and WS4)

for the maximum shear stress and the deformation capacity.

3 . 3 Deformation Capacity (WFl. WF2 , WFJl.

WFJ2. WFRl. WFR2. WFRC)

At first, flexural cracks appeared at the top and bottom. Thereafter, the crack width became

wider, and flexural shear and shear cracks at both top and bottom parts appeared and developed according to an increase in drift

angle. After a drift angle reached 1/200, ' diagonal shear cracks appeared in the specimens

except specimens WF2 and WFJ2; this diagonal

shear cracks became wider with an increase in the drift angle. The specimen WFR2 only failed

at the upper part. This failure was caused by

the incomplete filling of grout concrete at this portion. For specimens WF2 and WFJ2, compression failure and slippage occurred due

to development of cracks which appeared before

a drift angle of 1/200. The strains of main longitudinal and vertical

shear reinforcing bars of each specimen reached

0.2% through 2% at a drift angle of 1/400. Therefore, all specimens were supposed to be yielded in flexure before a drift angle of 1/400

.

Figure 8 shows the effect of shear reinforcement. From this figure, the difference in deformation capacity depending on amount of shear reinforcement is observed for a drift angle which is larger than 1/200. The specimen WF2 showed stable deformation behavior

without remarkable deterioration in strength

even at a drift angle of 1/50, although the specimen WFR2 didn't show better deformation

behavior at a drift angle of 1/50 than the specimen WFRl because of partial failure at the upper part

.

Figure 9 shows the effect of spiral reinforcement at critical compression zone. From the comparison of deformation behavior

after a drift angle of 1/100, it was observed

that the specimen WFJl showed a stable behavior

while the specimen WFl showed negative slope in positive loading. The load carrying capacity

of each specimen at a drift angle of 1/50 is in order of WFJ2>WF2>WFJ1>WF1 . This phenomena is explained by the reason why discrepancy of edge

zone due . to development of cracks and compressive failure might be prevented to some extent

by the spiral reinforcement.

Figure 10 shows the comparison regarding

splices in a wall panel. There is no remarkable difference in deformation capacity among

three specimens in the figure.

3.4 Effect of Transverse Wall (WS4 , WTTl . WS6

,

WTT2. WSRC. WTRC. WFl. WTCl

In the specimens WS4 and WSRC diagonal cracking

occurred at the drift angle of 1/500 and after

this cracking occurred the strength deterioration was severe. The specimens WTTl
and WTRC, which have a transverse wall at their

edge, reached its maximum strength at a drift

angle of 1/250 during positive loading, and

1/250 and 1/800 respectively during negative

loading. The strength deterioration in these

cases was medium, after reaching the maximum

shear strength (Fig. 11). The specimen WTT2 reached its maximum strength

at a drift angle of 1/130 in positive loading

and at 1/180 in negative loading. Its hysteresis loop was stable in comparing with that of

WS6 specimen (Fig. 12). WTCl specimen reached its maximum strength at a drift angle of 1/250 in positive loading, and

at 1/170 in negative loading. The mean shear

stress of the specimen WTCl was 1.4-1. 8 times

as much as that of the specimen WFl at those drift angles, but the hysteresis loops of the

two specimens were similar. Transverse wall of

the specimen WTCl inhibited the extent of the diagonal crack, but at the drift angle of

1/100, the transverse wall separated from the wall in the loading direction at its intersection (Fig. 13).

4. TEST RESULTS OF BEAMS

4 . 1 Effect of Shear Reinforcement

(GS2. GS3. GSRl. GSR2) Envelope curves of the mean shear strength

versus drift angle relation for specimens GS2

and GS3, GSRl and GSR2 are shown in Figs. 14 and 15. There are no difference between

specimens GS2 and GS3 for ultimate shear

strengths, but, after the drift angle exceeds

1/200, there is significant difference. The

two specimens showed diagonal tension cracking

which occurred at a drift angle from 1/300 to 1/200; there is no difference in the maximum

shear strength. The amount of shear reinforcement affected the behavior of the specimen

after the diagonal tension crack occurred. Specimen GS3, which has a large amount of shear reinforcement compared with the specimen GS2, maintained its maximum strength until a drift

angle reached 1/100. The specimen GS2 lost its strength immediately after the diagonal crack

occurred.

For GSRl and GSR2 specimens, a diagonal tension

crack occurred at a drift angle from 1/400 to 1/300. At this time, specimen GSRl reached its maximum shear strength, it couldn't maintain

that strength under a larger deflection. In specimen GSR2 , the shear strength increased

until a drift angle reached 1/50. After that, cracking occurred and compressive failure occurred where the critical compression was produced, which resulted in a reduction of shear strength.

4.2 Effect of R/C Slab (GSl. GS2. GFl . GF2

Envelope curves of the mean shear strength

versus drift angle relationships of specimens

GF2 (a beam with R/C slab) and GF3 (a rectangular beam) are shown in Fig. 16. Those specimens GSl, GFl (beams with R/C slab) and GS2 (a beam without R/C slab) are shown in Fig. 17. Specimens GF2 and GF3 showed flexural failure. Their maximum shear strengths were produced at

a drift angle in the range from 1/200 to 1/100. Those strength were 1.2 times as much as those estimated by the calculation where the slab reinforcement was accounted for GF2 . The specimens GF2 and GF3 maintained the strength

which is greater than the calculated strength

until the drift angle reached 1/50, and 1/100, respectively.

The specimens GSl and GS2 , a pair of short span

beams, reached their maximum shear strengths at drift angle of 1/250. Its strength was 1.38 to 1.52 times and 0.93 to 1.2 times larger than

the corresponding calculated strength. In the calculation of shear strength, the effect of R/C slab was not considered. Specimen GS2 could not sustain its maximum strength after a drift angle reached 1/250, and deterioration of

the strength was severe, while GSl maintained

the strength which is greater than the calcu- lated strength until a drift angle was reached

at 1/50. The stress of the inner (near the beam) reinforcing bar in the R/C slab reached yield

point under the loading cycle of 1/400, while

the outer one reached its yield point under

loading cycle of 1/200. In a short span beam

with R/C slab (GFl and GSl), all reinforcing

bars in the R/C slab reached tension yield

points under the large deflection; i.e. the drift angle exceeded 1/100.

5. CONCLUSIONS

1) The maximum shear stress increases with

increasing of axial stress.

2) The specimen failed in flexure has good deformation characteristics. Compared with

the flexural failure type specimen, the specimens failed in shear have 50% deformation capacity.

3) Increase in the amount of shear reinforcement (consequently increase in ratio of maximum shear strength to maximum

flexural strength) improves deformation

capacity of walls and beams.

4) The deformation capacity under a large amplitude of deformation is improved by the spiral reinforcement arranged at critical compression zone.

309
5) Splices of main longitudinal reinforcing

bars have little affect on deformation

capacity.

6) The larger the shear span ratio becomes,

the smaller the maximum shear stress becomes

.

7) The deformation capacity of the walls is improved with the existence of the transverse walls.

8) In rectangular masonry beams without R/C

slab, the ultimate strength can be esti- mated by the equation specified for a R/C

beam.

9) In the masonry beam with R/C slab, the shear strength is 1.45 times as much as the strength estimated by the equation speci- fied for R/C beams without considering the effect of R/C slab.

6. ACKNOWLEDGEMENTS

Many personnel in BRI have offered invaluable

help during the experiment. To all of them,

the authors wish to convey their gratitude,

especially to Dr. A. Baba, Messrs, M. Fujisawa,

H. Isoishi, and T. Kawashima. The authors also wish the express their sincere gratitude to Mr.

R. Kawai, Mr. T. Terada, Mr. M. Yamamoto, Mr.

A. Nakaoka, and Mr. Y. Yamaguchi who participated in the experiments as visiting research

members in BRI

.

7. REFERENCES

(1) S. Okamoto, Y. Yamazaki, A. Baba, M. Teshigawara, and H. Hiraishi, "U.S. -Japan Joint Earthquake Research Program

Involving Large-Scale Experiments on Masonry Building Structures - Japanese

Side Research Plan-," the 17th Joint

Meeting, U.S. -Japan Panel on Wind and

Seismic Effects, UJNR, Tsukuba, Japan, May

12-24, 1985.

(2) T. Okada, H. Hiraishi, T. Kamlnosono, and

M. Teshigawara, "Flexural Behavior of Reinforced Masonry Walls and Beams," The

3rd ASCE Engineering Mechanics Specialty

Conference, 1986.

(3) T. Kaminosono, H. Isoishi, Y. Yamaguchi,

and R. Kawai, "Seismic Capacity of Reinforced Masonry Walls Including Effects

of Axial Stress," The 4th Canadian Masonry

Symposium, Canada, 1986.

(4) M. Teshigawara and H. Isoishi, "Seismic Behaviour of Masonry Beams," The 4th Canadian Masonry Symposium, Canada, 1986.

(5) Papers for the First Joint Technical Coordinating Committee on Masonry Research

-U.S. -Japan Coordinated Earthquake Research Program-, Tokyo, Japan, August

26-27, 1985.

(5-1) H. Hiraishi, T. Kawashima, T. Terada, M. Yamamoto, and Y. Yamaguchi "Planar Test

on Flexural Behavior of Reinforced Masonry Walls".

(5-2) M. Fujisawa, "Seismic Behavior of Reinforced Masonry Walls - Effect of Shear Span Ratio-".

(5-3) M. Teshigawara and H. Isoishi, "Seismic Behavior of Masonry Beams".

(5-4) M. Teshigawara, H. Isoishi, and A. Nakaoka, "Effect of Transverse Wall Attached to Reinforced Masonry Wall".

310
Table 1 Material Properties

a) Prism Compressive Strength b) Reinforcing bars Specimen Prism Compressive

Strength (kg/cd) WS 1 1 8 3 WS 4 2 3 3 WS 7 1 8 3

i WSR 1 2 7 3

1 WS R 4 2 5 7

' WS R 7 2 1 8

! WSN 1 2 3 3

! WSN 2 2 3 3 WF 1 2 3 3 WF 2 2 1 9 WF J 1 2 1 9 WF J 2 2 1 9 WF LM 2 1 9 WF L 1 2 1 9 WFR 1 2 6 8 WFR 2 2 6 8 WTT 1 2 5 1 WTT2 2 5 1

1 WTC 1 2 5 1-

WS B 1 3 0 2 WS B 1 0 3 3 9 WSB 1 3 0 2 WSB 1 0 3 3 9 WSRC 2 7 3

1 WF RC 2 5 4

! WTRC 2 5 2

! GS 1 2 1 6 GS 2 2 1 2 GS 3 2 1 2

1 GS 4 2 1 2 GSR 1 2 7 5 G S R 2 2 7 5 GF 1 2 1 6

i GF 2 2 1 6 GF 3 2 1 6

Specimen <3 y (kg/cm*)

Flexural

re- bars

WSl, WS4.

WS7. WSRl.

WSR 4. WSR 7, WSN 1. WSN 2. WTT 1. WTT2.

WSB 1. WSB 1 0. WSRC. WTRC

3,948.0

WFl, WF2.

WF J 1. WF J 2. WFL 1. WFR 1. WFR 2. WTC 1. WFB 1. WFB 1 0. WFRC

3.542.0

WF LM 3,438.0 WSl. WS4.

WS7. WSR4.

WSR 7. WSN 1. WSN 2. WTT 1

.

WTT2. WS B 1. WSB 1 0. WSRC.

WTRC

1

1 3,791.0

Shear re- bars

(Vertical) WSR 1 3,708.0 WFl. WF2.

WF J 1. WF J 2. WFLM. WF L 1. WFR 1. WFR 2. WTC 1. WFB 1. WFB 1 0. WFRC

289.0

Shear re- bars

(Holizontal)

all ” W ” Specimens 3,620.0

Spiral WF J 1. WF J 2. WF L 1 3,193.0

Flexural

re- bars

GSl, GS2.

GS3. GS4.

GSR 1. GSR 2 3,833.0 GF 1 3.500.0 GF2. GF3 3,542.0

Shear re- bars

(Vertical)

GSl. GS 2. GS3. GS4.

GSR 1. GSR 2 3,791.0 GFl. GF2.

GF 3 3,289.0

Shear re- bars

(Holizonral)

GSl. GS2.

GS3. GS4.

GSR 1. GFl.

GF2. GF3

3,620.0 GSR 2 3,791.0

Slab re-bars GFl. GF 2. GSl 3.554.0

311
Table 2 Lists of Specimens

a) Wall Specimens

Pw TRANSVERSE''^

HALL

0 0138400

(0.1671) 0138200

(0.3341) 20138200

(0.6681)

M/QO At(cm*) X o X 0 X O X O

0.45

h/1.

180/200

016

(1.99) 2D19

(5.74)

HS1 .HSR1

2025

(10.1)

0.75

h/1.

180/120

016 (WS3.HSR3)

2019 teR2)

XIsji) (US2x2)

WS4 ,WS61 WSR4 ,HSN1 HSN2,WF1,HFR1 HFL1,HSRC,HFJ1 HFLH.WFRC HSB10<HSJ9WTT1

VfTCI WTRC

(WS5) (WSR5) HF2.HFB1 WFR2

HFJ2 WFB10

(WS6)

(HSR6)

WTT2

2025 (HS8) (HS11) (WS9) (MS 10)

1.13 h/1-

180/80

016 2019 HS7.HSR7

2025 VERTSCAL REINFORCEMENT DI69400

WS ; SHEAR FAILURE TYPE B1 ; 300nn LENGTH C/B SPECIMEN

HF ; FLEXURAL FAILURE TYPE RC ; R/C SPECIMEN

HT ; n;T-SHAPE, TC: CROSS-SHAPE SPECIMEN 0 ; OPEN JOINT SPECIMEN

R ; CLAY BLOCK N ; AXIAL FORCE SERIES SPECIMEN

J ; EFFECT OF LADDER STEEL

Tests of specimens in pharenses were conducted by other investigators

b) Beam Specimens

R/C SLAB

Pw 0139400 (20109400)“

0139400

0139200

0139150*'

20139200

0169150*'

M/QD At(cra^) O X O X o X

0.71 2016

(3.98)

GF1

l/h= 120/952019

(5.74) (GF6,GFR^

2025

(10.1)

GS1 GS2.GSR1

(gshi ,GSB^

(GSHZ)

GS3.GSR2

1.18 2016

1/hw 2019 GF2 GF3 (GFRI) (gF4,GFR^ (GF5,GFR|) 200/95 2025 GS4 HORIZONTAL REINFORCEMENT 0169400

FOR HOOP SERIES SPECIMEN

FOR 300nm LENGTH UNIT SPECIMEN

GS ; SHEAR FAILURE TYPE

GF ; FLEXURAL FAILURE TYPE B1 ; 300imi LENGTH C/B SPECIMEN

R ; CLAY BLOCK H ; HOOP SERIES SPECIMEN

Tests of specimens in pharenses were conducted by other investigators

312
Table 3

Tests

Results

1

Calculated

Results

I

- -

i96 0

ISOI 1.187 1.236 0.975 1.028 [

0.9861

1.071 0.636 0.580 0.620 10.576 1.150 1.081 1.151 1.125

I

1

si 866 0

996 0 !!

1.109 1.012 1

177

0.909 0.966 1.025 1.033 1

0.880 9«I I

ROR 0

1.109 1

1.047_|

0.636 0.585 0.908 0.945 0.597 0.649 1.057 0.965 m iii

t

1 -

i

1

0.773 ~j

0.597 0.710 0.753 1.041 0.886 0.615 0.902 1.039 0.980

(19.17) 24.85 (24.52) 31.39 (17.96) 23.64 (19.20) 24.88 25.50 20.30 is.

!

i

1

1

1 P

1

1 ®

1

!

19'RI 1

20.81 17.13 21.56

!

-

(26.67) 41.05 (26.67) 41.05 (16.15) 27.03 (26.68) 41.06 I

15.34

i s s

(8.90) 15.33 1

a (28.34) 39.40 a .

j

i ,a

17.01 ! ^

: ^

28.34

S

Q£

6.630 9.990 If

9.300 If7.150 tt

11.750 11.380 21.063 4.369 r

9.750 7.880 8.130

10.750 31.400 19.400 21.600 18.100 22.800 12.000 11.200 23.500 000'9 OOTi11.500

10.2509 4.800 5.900

OOS'9

OOTOI

4000 01

006 01

tr

3.458 5.080 7.740 5.500 4.120 5.810 4.494 1.311 2.320 2.390

10.120 5.240 4.240 2.140 9.680 4.960 osrs OUS 9.210 5.110 4.680 5.430 4.380 3.463 5.060 9.700 4.620 4.500 6.830 4.330 7.630 8.270

}

26.12 24.02 37.27 38.81 23.05 24.31 S3

sa

15.56 14.20 15.80 14.69 23.34 21.94 38

SS3

25.41 23.98 18.49 19. 10

11.64 10.47 27.96 25.52 21.90 16.91 20.11 21.33 17.70 15.07

I as

IJ

1 + 1 + 1 + 1 + 1 + 1 + 1 + 1 1 ! + .

! - '

1 - i I i S 1

NFBIO IHSBIO

i s 5 s g 2

1 i

1

Calculated

Results

|

- W 1

1.113 1.040 1.235 1.313 Si 1.213 1.218 [

1.134 1.272 m

j oo

0.752 0.713 1

0.933 ]

0.939 0.733 0.726 f0.946 0.942

1l|| 0.939 0.962

I eI

1 oo

0.979 0.921

!-=

0.806 0.628 0.781 0.730 1.007 1.070 0.780 0.812 0.871 0.875

0.970 1

1.087 pi

oo

0.598 0.615 10.794 0.872 ii

— o

1.046 0.991 1^018 1.025 1

si 8^

-iO

1.046 0.973 0.993 1.019 0.936 1.124 1.094 1.029

-

16.58 18.66 16.36 17.49 19.46 117.30 20.27 21.88 19.79 17.45 22.43 18.26 23.56 17.08 17.08 S 22.43 18.05

38.11 a Ss

40.07 27.09

20.23 1

136^41 11

43.96 I

26.76 2 i s S 3 3 3 a

1

tr

2.220 1.045 3.778 5.833 8.205 9.017 1.769 1.500 6.625 5.250 6.051 3.830 4.778 4.222 4.556 4.000 5.278 2.778

12.625 9.375 20.500

If

20.5002 7.679 8.715

10.938 12.177 16.000 11.375 18.561 11.250

20.0009 15.811 II 17.750 11.750

cs

1.720 1.211 9.359 8.923 2.010 1.806 2.510 2.280 2.111 2.345 4.142 2.503

10.381 2.689 4.926 2.415 4.963 2.480 5.092 2.567 5.040 5.040 5.072 5.44219.607 10.839 9.483 5.047

-

30.73 23.92 20.77 19.40 20.21 21.48 31.25 32.55 SS

S3 S3 S8 25.95

23.10 1 as

as: 383S3

16.27 15.96 16.88 15.99 17.27 17.10 16.16 16.10 16.03 16.44 15.10 18.15 So

Speciaen + 1 1 + 1 1 1 + 1 + 1 ^ 1 1 - 1 + 1 + 1 + 1 1 ^ 1 $ S S

1 i s 1

Si% i i i i %

313
X : maximum shear stress obtained from test (kg/cm^)

X : drift angle at r (lO'^rad.)

8 : dri f t angle at 0.8 • ^ m»x (lO'^rad.) ^ ( 0.053Pto®-"Mfm’fl80)/(M/QO+0.12)+2.7/7::;p:r+O.Uoe} B^j/BD - - - eq .1 where M/QD : shear span ratio I^M/QD^3

D : wall length or beam depth (cm) d : effective wall length or effective beam depth (cm)

j : 7d/8 (cm)

B : wail width or beam width (cm) Be : equivalent width of wall or beam (cm) Pie: flexural reinforcement ratio lOOxat/Bed (%) fm’ : prism compressive strength (cm) yield strength of shear reinforcement (kg/cm^)

Pwo • ratio of shear reinforcement

'^oe: vertical compressive stress (kg/cm^)

(from A.l.J R/C Design Standards)

r bu-2 • 0.9 a,(?yd /am

rb„-2 { 0.9 at.? yd 0.4a«<rw+0.5ND(l-N/BDf„’)) /hBD

‘ where at : cross sectional area of flexural reinforcement (cm^)

<7 y : yield strength of main vertical reinforcement (kg/cm^) 3w : cross sectional area of vertical reinforcement (cm*)

: yield strength of vertical reinforcement (kg/cm*)

N : vertical compressive load (kg)

h : clear height of wall (cm) £ : clear length of beam (cm) (from A.l.J R/C Design Standards)

Fig. 1 Unit

a) Standard Concrete Block Unit

30 12D

1 ^ 1 ^

b) Standard Clay Brick Unit

314
Fig. 3 Loading System

315
...M r ;\V^.>-U 1 ...

.. 'K '

1 V'

\\ykj. :_._K l\ 1

1 / I

1 /! 1 y 7 \

1 :v 57^1 ./1

a) WS4 b) WFJ2

Sj\ " ^Vl

Vm^f\ t )(

7^

.{ II Tfrr AfMv T7V-^ 1/ . 3Sj^

e) GF2

Fig. 4 Crack Pattern

f) GF3

Fig. 5 Effect of Axial Stress

316
r(k«/c.3) T(k«/c.^)

Fig. 6 Effect of Shear Span Ratio (Concrete Block)

Fig. 7 Effect of Shear Span Ratio

(Clay Brick)

Fig. 8 Effect of Shear Reinforcement Fig. 9 Effect of Spiral Reinforcement

317
T(ta/orf)

Fig. 11 Effect of Transverse Wall

Fig. 10 Effect of Splice (T Type Pw=0.167%)

Fig. 12 Effect of Transverse Wall

(T Type Pw=0.667%)

Fig. 13 Effect of Transverse Wall

(+ Type Pw=0.167%)

318
Fig. 14 Effect of Shear Reinforcement

(Concrete Block Beam)

Fig. 15 Effect of Shear Reinforcement

(Clay Brick Beam)

Fig. 16 Effect of R/C Slab

(Flexural Type)

Fig. 17 Effect of R/C Slab

(Shear Type)

319
U.S.-JAPAN COORDINATED PROGRAM ON MASONRY RESEARCH

by

1 2 Shin Okamoto , James Noland , M.ASCE

ABSTRACT

An overview of the U.S. -Japan Coordinated Program

on Masonry Research is presented. The program

consists of parallel efforts in Japan and the United States to develop design methods for reinforced masonry buildings located in areas of seismic risk. Both programs rely heavily upon experimental work to produce data and confirm

design methods. Both programs also include tests

of a full-scale reinforced masonry building

specimen to observe and verify system response.

1. INTRODUCTION

The U.S.-Japan Coordinated Program on Masonry

Research is the third in a series of building research programs conducted by Japan and the United Stated under the Auspices of the Panel on Wind and Seismic Effects of the U.S.-Japan Natural

resources development program (UJNR) . The first

two were conducted on reinforced concrete and structural steel buildings.

The joint masonry programs are intended to obtain experimental data and establish design methods

pertinent to design and construction of masonry

buildings in areas of seismic risk. However,

the knowledge and techniques developed in the course of the two programs will also be applicable to design and construction of masonry

buildings subjected to other types of loading.

The two research programs have begun. The pro- gram in Japan began in 1984 and the U.S. program

commenced in 1985. This paper presents a brief overview of each program.

2. PART I - OVERVIEW OF THE JAPANESE PROGRAM

2. 1 Background Masonry structural systems are known worldwide

as traditional and efficient with high durability,

fire resistance, sound and heat insulation, and high flexibility in construction. The structural

system was imported to Japan from Europe with

little attention given to earthquake resistance

of masonry buildings as then constructed. The masonry buildings constructed by the imported

concepts sustained extensive damage in the Great

Kanto Earthquake. Since then the construction of masonry structures practically ceased in Japan

and little effort was made on the study of the seismic resistance of masonry structural systems. Because of the large demand- for building construction after World War II, Japan developed various construction systems amenable to mass production

of buildings. Masonry buildings were generally

not built because of the problems then perceived

regarding masonry structures, e.g., tedious labor. What is desired in building construction today

has changed. Now more value is placed on quality

then quantity. Versatility and variety are more appreciated than uniformity.

It was natural, therefore, to reconsider masonry

structures. Since the late 1960's various types

of new masonry systems have been developed to make construction more reliable and competitive. The reinforced concrete block (RGB) system is one example. The Building Research Institute has contributed

a great deal to the improvement of masonry structural system. One current project is entitled "Application of Small PC units to Urban Housing"

in which a high quality masonry unit will be developed and a new earthquake resistant construction technique proposed. Additional research is required to further improve masonry design and construction techniques for greater seismic safety thus enabling

greater utilization of this type of construction.

The immediate target of the Japanese part of the U.S.-Japan joint research program will be the development of a masonry construction system applicable to low and medium rise building structures . 2.2 Research Team Organization

The organization of the Japanese masonry research

effort is shown in Figure 1 and its relationship

to the U.S. masonry research team organization.

The U.S. organization is presented in more detail

in Part 2 of this paper. Note that the Japanese

and U.S. research efforts are linked at the government level by a coordinated research agreement and at the project level by the Joint Tech- nical Coordinating Committee on Masonry Research

(JTCCMAR)

.

Three major committees are included in the Japanese masonry research organization, i.e., the Technical Coordinating Committee on Masonry

Research (PROCMAR) , and the Building Construction Committee on Masonry Research (BLDCMAR) . The

Director, Production Department, Building Research Institute Ministry of Construction,

Tsukuba, Japan. Principal, A tkinson-Noland & Associates.

Boulder, Colorado, U.S. A.

320
first committee, TECCMAR, coordinates technical research planning on materials, components, assemblies and full-scale experiments and will

develop proposals for aseismic design guidelines

for low and medium use reinforced masonry building structures in Japan. PROCMAR was organized under a domestic cooperative research agreement between the Building

Research Institute, the Japan Association for the purpose of promoting the U.S. -Japan research

program in Japan on masonry structures. The members of PROCMAR include government officials,

local self-governing body officials and repre- sentatives from many industrial organizations as shown in Figure 1. There are many construction problems to solve in this research program to develop technology for

the design and construction of low and medium

rise reinforced buildings in, Japan. The third committee, BLDCMAR, was organized under PROCMAR

to address the problems concerning reinforced masonry building construction techniques.

2.3 Research Plan

The Japanese masonry research effort will take approximately five years and is composed of the following basic components:

1. Material Tests

2. Static Tests of Walls

3. Static Tests of Beams

4. Static Tests of Wall and Beam Assemblies

5. Full-Scale Planar Static Tests

6. Shaking Table Tests of a Scale Model

Specimen

7. A Full-Scale Five-Story Structure Test

8. Establishment of Aseismic Design Guidelines

The time schedule for the Japanese research

effort is shown in Figure 2. During the first

two years, fundamental tests of materials, of

vrall and beam components, masonry structural

assemblies, and of a full-scale planar specimen

are planned. A great deal of this has been done. The shaking table tests of a scale model specimen

and a full-scale five-story structure will be done in the later part of the program.

2.4 Japanese Research Objectives

The final target for the Japanese side of the joint research program is to propose aseismic

design guidelines for low and medium rise reinforced masonry building structures in Japan. Presently, only reinforced masonry buildings of

12 meters (one to three stories) are permitted.

As a result of this program design guidelines for low-to-medium rise reinforced masonry building

(up to five stories) will be proposed.

There are also many other objectives related to the structural design of masonry buildings such

as

1. To assure acceptable aseismic performance. Aseismic performance which is substantially equal to that to be required for wall type RC buildings in Japan will be expected for reinforced masonry buildings.

2. To propose decreased wall length rate.

2 Revise 21 cm/m in the present regulations : 2 a) 15-18 cm/m for five story buildings

and,

2 b) 12-15 cm/m for three story buildings.

3. To eliminate the use of reinforced con- crete collar beams. Reinforced concrete collar beams are required in reinforced masonry buildings under present regulations.

4. To simplify joint works of reinforced

bars. Effectiveness of lap and mechanical joints of vertical reinforcing bars

in walls must be established.

2.

5

Review of Research Done in 1984 The following topics were addressed in 1984:

1. Masonry Units - Standard concrete and clay block units developed for use in the Japanese program are shown in Figures 3 and 4.

2. The predictions of the compressive

strength of prisms based on the properties of block units, grout concrete and mortar has been investigated.

3. High lift grouting procedures have been

studied using wall specimens 2h units

long (1000 mm or 39.4 in.) by 12 courses high (2400 mm or 94.5 in.).

4. Effectiveness of lap splices of rein- forcing bars in masonry has been studied using six types of splice configuration.

This included evaluation of spiral rein- forcement around lap splices.

5. Anchorage of reinforcing bars in grouted masonry has been studied considering

grout cover and proximity of the rein- forcement to face shells of units.

6. Static tests of masonry wall and beam components were conducted. The major purpose was to investigate their defor- mation performance and shear capacity

varying area of main reinforcement, shear reinforcement and shear-span ratio.

321
3. CONCLUSIONS Masonry buildings of either concrete or clay unit masonry are at present few in Japan compared with

the number of buildings of other materials, e.g., wood, concrete, and steel. This is because of

the susceptibility to earthquake damage of masonry buildings previously constructed by methods imported from Europe one hundred years

ago. It is expected that the five-year masonry

research program just described on low-to-medium

rise masonry buildings will allow and encourage

greater numbers of such buildings in Japan.

4. ACKNOWLEDGEMENT

The author of Part 1 of this paper wishes to express his sincere thanks to TECCMAR-Japan

members who contributed great effort to the tech- nical research planning of the Japanese program

and for coordination with the U.S. program. Gratitude is also expressed to PROCMAR members

for their contributions to promote the Japanese

side of the joint U.S. -Japan masonry research

program.

5. PART II - OVERVIEW OF THE UNITED STATES PROGRAM

5. 1 Background

Load-bearing masonry buildings have been built

in the United States for many years—nearly since

the time European settlers first arrived. Masonry buildings are a significant percentage

of buildings built from that time up to the

1930' s and are generally unreinforced. Reinforced concrete and structural steel buildings gradually became a larger and larger part

of the construction market. Their use was and

is encouraged by the great amount of research

and development done and by the gradual improvement in design methods and codes.

In contrast, the use of masonry as the primary

structural system declined in comparison to the use of steel and concrete possibly due to the perception that masonry design technology was

not at the same level as that for steel and con- crete and to the perception that masonry structures perform poorly in earthquakes. While some U.S. research in structural masonry

was done earlier, it beban to increase in the mid-to-late sixties and continues to increase. Masonry building code improvement has also recently been emphasized as exemplified by the 1985 Uniform Building Code and by the work of the joint ASCE/ACI Committee 530 on masonry. This author believes that some of the impetus for renewed interest in the structural use of masonry

is due to its architectural flexibility and that

it can be competitively economic for certain

types of building plans. The introduction of reinforcement has been shown to provide the ductility required for adequate performance in earthquake conditions.

However, the design methodology is still basic- ally a working stress approach based on the assumption of linear-elastic material behavior although the latest UBC does permit strength

methods for certain applications. While many

reinforced masonry structures have been built and

have performed successfully, economic considerations and a better ability to predict ultimate behavior would make structural masonry a more viable alternative as a building system particularly for earthquake conditions. A need was recognized for complete strength-method design

techniques for reinforced masonry based upon adequate experimental data to bring masonry

structural technology to a level consistent with modern needs.

5. 2 Research Team Organization

A team of masonry researchers known as the Tech- nical Coordinating Committee for Masonry Research/

U.S. (TCCMAR/U. S . ) was formed to identify the research necessary and to organize in a manner

such that the research can be done in an orderly

fashion. The organization of TCCMAR/U. S. and the basic research areas are depicted in Figure 1. Funding is provided by The National Science Foun- dation (Dr. A.J. Eggenberger) and liason with the UJNR Committee on Wind and Seismic Effects is done by G.R. Fuller (HUD) and Dr. H.S. Lew (NBS). A single coordinator (J. Noland) is responsible

for overall operation of the U.S. program with

the advice and assistance of the Executive Panel. The Consultants Panel is composed of individuals highly qualified to provide basic guidance on the course and conduct of the program. Two were intimately associated with previous U.S. -Japan

joint research programs. A panel of Industry Observers was formed to pro- vide industry perspective and to relay information about the program to the industry. The pur- pose of the Industry Participation Panel is to define ways in which the industry can assist the program and arrange for that assistance.

5. 3 Research Plan

The U.S. masonry research program consists of a group of coordinated research tasks beginning with tasks which address basic material behavior, progressing to tasks addressing behavior of masonry structural components, e.g., walls, and finally to experiments on a full-size segment of

a masonry building. Development of analytical methods for stress and strain analysis and system behavior will proceed in parallel with experi- mental tasks. The final task will be the preparation of recommendations for design procedures and building code provisions.

The research tasks which were identified by TCCMAR/U. S. fell into 10 categories as shown in Figure 1. The following is a more detailed list

of research tasks in each category and their

obj ectives.

322
Category Task Title-Purpose

1.0 1.1

1.0 1.2

2.0 2.1

2.0 2.2

2.0 2.3

Preliminary Material Studies - To establsih the range of con- tinuity of masonry behavior to provide a basis for selection

of the type or types of masonry to be used. To establish standardized materials test procedures for all the experimental tasks. Material Models - To evaluate

Kl, K2, and K3 for the flexural

stress-block. To determine

uniaxial and biaxial material

properties for analytical

models (Tasks 2.1 and 2.2) including post-peak behavior.

To evaluate non-isotropic

behavior. Force-Displacement Models for Masonry Component - To develop force-displacement mathematical

models which accurately characterize reinforced masonry com- ponents under cyclic loading

to permit pretest predictions

for experimental results. To develop models suitable for parameter studies and models

suitable for design engineering. Strain Analysis Model for Masonry Components - To develop

a strain model for reinforced masonry components in conjunction with Task 2.1 to enable

regions of large strain to be identified thus assisting in experimental instrumentation

planning. To develop a sim- plified model to be used to provide data for strength

design rules and in-plane shear design procedures.

Dynamic Response of Masonry

Buildings - To develop a gen- eralized dynamic response model

to predict inter-story dis- placements using specified time histories. To correlate force displacement models and to investigate force-displacement characteristics of structural

components in the near-elastic

and inelastic displacement

range. To provide data for building test planning.

Dynamic Properties of Masonry

Systems - To develop consistent unified, rationale for seismic design of masonry buildings

3.0 3.1(a)

3.0 3.1(b)

3.0 3.2(a)

3.0 3.2(b)

4.0 4.1,

considering elastic and in- elastic response of masonry

buildings and of the soil/ structure interaction and related to seismic hazard

zones. Response of Reinforced Masonry

Story-Height Walls to Fully Reversed In-Plane Lateral

Loads - To establish the behavior of story-height walls

subjected to small and large amplitude reversals of in-plane lateral deflection, axial

force, and bending moments considering aspect ratios, reinforcement ratios and patterns, and the effect of openings.

Development of a Sequential

Displacement Analytical and Experimental Methodology for

the Response of Multi-Story

Walls to In-Plane Loads - To develop a reliable methodology

for investigating, through integrated analytical and experimental studies, the in- plane behavior of multistory

reinforced hollow unit masonry

walls. The methodology will

be the basis of studying the response of a full-scale masonry research building. Response of Reinforced Masonry

Walls to Out-of-Plane Static Loads - To verify the behavior

of flexural models developed

using material models, to evaluate the influence of mortar joint spacing, unit properties, reinforcement

ratios and grouting upon wall

behavior. To provide stiffness data for correlation with

dynamic wall test results

(Task 3-2 (b)). Response of Reinforced Masonry

Walls to Out-of-Plane Dynamic Excitation - To determine effects of slenderness, rein- forcement amounts and ratios, vertical load and grouting on dynamic response, to verify mathematical response models,

to develop design coefficients

for equivalent static load methods

.

Wall-to-Wall Intersections and Floor-to-Wall Intersections

of Masonry Buildings - To determine the effectiveness of

323
4.2

5.0 5.1

5.0 5.2

6.0 6.1

6.0 6.2

intersection details to connect masonry wall components, to construct a nonphenomenological analytical model

of intersection behavior.

8.0 8.2

9.0 9.1 Assembly of Existing Diaphragm

Data - To assemble extensive

existing experimental data on various types of floor dia- phragms, to reduce to a form required for static and dyna- mic analysis models. Grouting Procedures for Hollow

Unit Masonry - To identify methods of grouting hollow

unit masonry such that the cavity is solidly filled and reinforcement is completely

bonded.

Concrete Plank Diaphragm

Characteristics - To investigate experimentally concrete plank diaphragm floor dia- phragms to determine modes of failure and stiffness characteristics including yielding

capacity in terms of distortion as needed for masonry

building models.

Reinforcement Bond and Splices

in Grouted Hollow Unit Masonry 9.0 9.2 To develop data and behavioral models on the bond strength and slip characteristics of deformed bars in grouted hollow unit masonry,

to develop data and behavioral 9.0 9.3 models on the bond strength

and slip characteristics of deformed bar lap splices in grouted hollow unit masonry

as needed for building modeling.

reduction factors. To review program experimental research

tasks assure that statistical

benefits are maximized and proper limit states are investigated. Numerical Reliability Indices

- To develop numerical values

of statistically-based strength reduction (i.e., 0) factors using program experimentally

developed data, other appli- cable data, and judgement.

Design of Reinforced Masonry Research Building, Phase I - To develop the preliminary

designs of the potential research buildings which

reflect a significant portion

of modern U.S. masonry con- struction. To estimate inter- story displacements using methods developed in Category

2 tasks and the associated

load magnitudes and distributions. To select a single

configuration in consultation with TCCMAR which will be used

as a basis for defining equip- ment and other laboratory facilities in Task 9.2. Facility Preparation - Define, acquire, install and check-out equipment required for experi- ments on a full-scale masonry

research building. Full-Scale Masonry Research

Building Test Plan - To develop a detailed and com- prehensive plan for conducting

static load-reversal tests on

a full-scale reinforced masonry research building.

7.0 7.1 Small Scale Models - To experi- mentally evaluate the use of small-scale modeling for reinforced hollow-unit masonry

walls by correlating test results with test results of full-scale walls of the same configuration. To determine

if tests of small-scale specimens can reveal basic characteristics and failure modes of full-scale masonry specimens.

8.0 8.1 Limit State Design Methodology

for Reinforced Masonry - To select an appropriate limit

state design methodology for masonry. To select and document a procedure to compute numerical values for strength

9.0 9.4 Full-Scale Test - To conduct experiments on a full-scale reinforced masonry research building in accordance with

the test plan and acquiring

data indicated. To observe building response and adjust

test procedures and data mea- surements as required to establish building behavior.

10.0 10.1 Design Recommendations and Criteria Development - To develop and document recom- mendations for the design of reinforced masonry building

subject to seismic excitation

in a manner conducive to design office utilization. To develop and document cor324
responding recommendations for masonry structural code provisions . 11.0 11.1 Coordination - To fully coor- dinate the U.S. research tasks

to enhance data transfer among researchers and timely completion of tasks. To schedule

and organize TCCMR and Executive Panel meetings. To establish additional program

policies as the need arises.

To stimulate release of progress reports and dissemination

of results. To coordinate with

industry for the purposes of informing industry and arranging industry support. To interface with NSF and UJNR

on overall funding and policy

matters.

The names, affiliations and research task arrangements of current TCCMAR/U.S. members are as follows

Task Researcher Institution

1.1 R. Atkinson Atkinson-Noland & Assoc.

1.2(a) A. Hamid Drexel Univ.

1.2(b) R. Brown Clemson Univ.

2. 1 R. Englekirk AKEH Joint Venture

2.2 G. Hart " " " 2.3 J. Kariotis II II II 2.4 R. Ewing II 3.1(a) J. Noland/

B. Shing

Univ. of Colo. -Boulder

3.1(b) G. Hegemier/ Univ. of Calif. -San

F. Seible Diego

3.2(a) A. Hamid Drexel Univ. 3.2(bl) S . Adham Agbabian Assoc.

3.2(b2) R. Mayes Computech

4. 1 G. Hegemier Univ. of Calif. -San

4.2 F. Seible Diego

5.1 M. Porter Iowa State Univ.

5.2 " II II II 6. 1 L. Tulin Univ. of Colo . -Boulder

6.2 " II II II 7.1 A. Hamid Drexel Univ.

8. 1 G. Hart Univ. of Calif. -Los Angeles

8.2 TCCMAR* —

9. 1 J. Kariotis Barnes-Kariotis

A. Johnson Joint Venture

9.2 G. Hegemier Univ. of Calif. -San

Diego

9.3 TCCMAR* —

9.4 TCCMAR* Univ. of Calif. -San Diego

10. 1 TCCMAR* —

11. 1 J. Noland Atkinson-Noland & Assoc. Task to be done by TCCMAR/U.S. or subgroup

thereof.

5 . 4 Task Coordination

The research tasks are interdependent, i.e., results from a given task may be required for the execution of others and vice-versa. Analytical

tasks will generally require interaction with experimental tasks on a fairly continuous basis

so that analytical model development may incor- porate data as they are obtained. The needs of

the analytical tasks will in turn serve to define,

in part, the manner in which experimental tasks are designed and conducted and the data to be obtained.

The anticipated intra-task interaction is depicted generally in Figure 2.

5. 5 Schedule

The schedule for tasks comprising the U.S. pro- gram is shown in Figure 3. Total time required

to complete the program is estimated to be approximately six years.

5.6 Review of 1985 Progress

Task 1.1, Preliminary Studies was completed by June 1985. Work was begun on Tasks 1.2(a & b)

,

2.1, 2.2, 2.3, 3.1(a), 5.1, 5.2, 6.2, 8.1 and

9.1 in the Fall of 1985; therefore results are not yet available

6. CONCLUSIONS

The U.S. program was designed to provide a strength design methodology for reinforced masonry buildings based upon an integrated experi- mental and analytical program of specific research

tasks. Study of the task descriptions will reveal that only issues believed to be essential

to this development are addressed. Budget limi- tations also restrict the number of parameters which can be evaluated in these tasks. However,

the work will provide a unified framework and reference point so that future research, as funds become available, will be able to effectively

build upon the knowledge developed in this program.

7. ACKNOWLEDGEMENT

The support of the National Science Foundation is greatly acknowledged. The contribution of hollow concrete units by the Concrete Masonry Association

of Calif ornia-Nevada and of hollow clay units by

the Western States Clay Products Association for specimen construction is also acknowledged.

325
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U.S ° JAPAN COORDINATED PROGRAM FOR MASONRY BUILDING RESEARCH RGURE 1 = TCCMAR ORGANIZATION (U.S.)

U.S.-JAPAN COORDINATED PROGRAM FOR MASONRY BUILDING RESEARCH

FIGURE 2 — TASK INGREDIENTS-DEPENDENCE CHART (U.S. PROGRAM)

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326
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327
SEISMIC BEHAVIOR OF A FULL-SCALE K-BRACE STEEL BUILDING STRUCTURE

BY Hiroyuki Yamanouchi*, Mitsiamasa Midorikawa*, Isao Nishiyama* and Tatsuo Murota*

ABSTRACT

Full-scale seismic tests on a steel building

were performed as a three-year research project

from 1981. This research is part of the U.S.- Japan Cooperative Earthquake Research Program Utilizing Large-Scale Testing Facilities, to compare full-scale building behavior with scale model behavior and to assess the damage and

safety levels of buildings subjected to earthquakes designed by current design

practices

.

This paper reports the results of seismic tests

on a full-scale six-story steel building with

concentric -K braces. A six- story, 2x2 bay, steel building was constructed at the Large

Size Structure Laboratory of the Building

Research Institute of the Japanese Ministry of Construction. The seismic tests were run as a six degree of freedom pseudo -dynamic system

(on-line computer system). The input excitation was the NS component of the 1978 MiyagiKen-Oki Earthquake; and maximum intensity was

set at three levels. To simulate working load

conditions, the earthquake motion was scaled to

65 gal; for a moderate earthquake the peak

intensity was set at 250 gal; the maximxim earthquake was run at 500 gal. The test results showed large capacity of the pseudo -dynamic testing technique, the impor- tance of precise prediction of the interaction

of the post-buckling behavior, local failure of

the braces, the elasto-plastic behavior of the moment-resisting frame, and that of the three dimensional effects of the overall test building

.

1.

INTRODUCTION

As a part of the U.S. -Japan Cooperative Research Program Utilizing Large-Scale Testing

Facilities (Ref. 7), a six-story, 2 by 2 bay, concentrically K-braced steel building was constructed in full-scale, and subjected to Miyagi-Ken-Oki simulated earthquake of different intensities by the pseudo -dynamic (PSD) test facilities (on-line computer-actuator

test). To simulate working load conditions,

the earthquake motion was scaled to 65 gal; for

a moderate earthquake the peak intensity was

set at 250 gal; the maximum earthquake was run

at 500 gal. In a series of these seismic

tests, the overall inelastic behavior and the

*Building Research Institute, Ministry of Construction

local behavior of the three dimensional full- scale structure were obtained, and considerations were carried out from various viewpoints

.

2. FULL-SCALE TEST BUILDING

The test building was designed to satisfy the requirements of the 1976 Uniform Building Code and the 1981 Seismic Design Code of Japan. The plan and elevation are shown in Fig. 1. This building structure consisted of two unbraced moment-resisting frames (Frame-A and -C) and a concentrically K-braced frame (Frame-B) . The dead load of the building was 523.6 tons (1154

kips) while the design dead load was 634.7 tons

(1399 kips). Member size of the test structure is shown in Table 1. Principal design criteria and details

are; 1) The design base shear coefficient

should be 0.197. 2) Live load and exterior

wall weight should not be included in the design earthquake lateral forces. 3) Columns

and girders should be wide flange shapes of ASTM A36 structural steel, and braces be square tubing section of ASTM A500 Grade B structural

steel. 4) Girders and floor beams should act compositely with the floor using shear studs.

5) Braces should resist both tension and

compression. 6) Girder-to-column connections

should be moment connections in loading

direction. 7) Column bases should be fixed.

8) Braces should be directly welded to surrounding frames without gusset plates.

3. TEST PROGRAM AND PROCEDURE

All tests, exclusive of free and forced vibration tests, were run as a six degree of freedom pseudo -dynamic (PSD) system. The input excitation was the NS component of the 1978 Miyagi-Ken-Oki Earthquake (Tohoku University)

.

To maximize the amount of knowledge gained from

the program, the test sequence in Table 2 was

adopted, including the three level major

seismic tests (the elastic PSD, inelastic PSD Moderate and inelastic PSD Final tests). At

the head and end of the test sequence, the free and forced vibration tests were performed to obtain modal parameters and damping

characteristics of the test structure.

The test structure was loaded by eight servo- controlled actuators through rigid loading

beams installed at the edge of each floor. The

328
roof floor had two actuators, with a capacity

of ± 100 ton (220 kips) load and ± 100 cm (39.4

in.) stroke, to prevent the test structure from overall torsion by displacements from the two jacks. At the 6th floor, two jacks were arranged with a capacity of ± 100 ton and ± 50 cm (19.7 in.). Only one of the two jacks were

used in the PSD test control. The other

actuator was controlled to generate the same load as the PSD-controlled actuator. Excluding

the upper two floor levels , one actuator with

the same capacity as the actuators at the 6th

floor was set up for each floor. The PSD tests were run at time intervals of 0.01 sec. for the elastic level and 0.005 sec. for the moderate

and final tests, respectively.

The basic instrumentation plan consisted of two phases. The first measured floor- level displacements and actuator forces for the PSD

tests at each time step, by using digital-type

displacement transducers and load cells. The second detected deformations of members of the steel skeleton, such as rotation angles of

members, panel zone distortions of girder- to- column connections, and strains on structure components as girders, columns, braces and reinforcing bars in slabs, to estimate the

load- deformation characteristics of the structural members of the building. Approximately 1,000 channels of data were collected

for this purpose; the data were sampled at every fourth time step during the major three

PSD tests.

4. OVERALL RESPONSE AND GROSS BEHAVIOR OF TEST STRUCTURE

Response waveforms of floor level displacements

as a function of time, from the moderate and

final tests are shown as solid lines in Figs. 2 and 3. In the moderate test, the girder panel

zone at the brace junction in the middle of the span failed at the 2nd floor. Also the braces

at the 2nd, 3rd and 5th stories buckled

slightly. The girder panel -zone was repaired

by placing it with a new plate before the final

test

.

In the final test, at the 1st through 5th

stories, the braces buckled in-plane and/or

out-of-plane of the frame. In particular, the braces at the 2nd and 3rd stories buckled

laterally 21 to 24 cm (8.27 to 9.54 in.) with

local buckling and tearing at the mid- span and both ends. Many panel-zones of girder- to- column connections and member ends also

suffered, mainly in the lower three stories.

It should be noted that the maximum roof displacement reached was 22 cm (8.66 in.) and

the maximum base shear was 331 tons (729 kips)

at a displacement of 2.3 cm (0.91 in.) in the

1st story. Fig. 4 shows the relationship of story shear force vs. interstory displacement

in the final test (solid lines). In this

figure, it is shown that the lower stories

reached their load carrying capacity. On reaching this force level, several braces

degraded. The vertical displacement at the brace junction reached 3.5 cm (1.38 in.) at the 3rd floor by a significant loss in capacity of

the braces. Finally the test structure

sustained an interstory drift angle of 1/57 at

the 2nd story, and the north side brace at the

3rd story ruptured completely.

In Figs. 2, 3 and 4, analytical results are also shown in dotted lines. The DRAIN-2D

computer program developed by Kanaan and Powell

(Ref. 5) was used for the analysis. In this analysis, a degrading hysteresis model (Fig. 5) was applied to the post -buckling and cyclic behavior of braces. This model was developed

by modifying the Jain-Goel-Hanson hysteresis

rule of braces (Ref.. 3). The comparison

indicates that the analytical model is very

good in predicting the overall behavior of the test building.

5. LATERAL ELASTIC STIFFNESS OF EACH STORY

Lateral elastic stiffness of each story is defined as a value of story shear divided by

interstory drift. By the Elastic test, the stiffness of each story was obtained as listed

in Table 3. On the contrary, an analytical

stiffness of each story was obtained by using a computer program for static analysis on the basis of the DRAIN- 2D. However, the results of

this analysis (Analysis 0) had significant

deviations from the test results as shown in

Fig. 6. This discrepancy would be due to 1) three dimensional (3-D) effects of the framing

of the test structure and 2) inaccurate estimation of each member's stiffness such as the stiffness of composite girders.

To study the effects of the above factors on

the story stiffness in the Elastic test, further analysis were performed by taking the following items into account:

1) The nodal horizontal displacements at the nodes of B1 and the mid- span of the braced

girders must be independent of those at the other nodes. Also reasonable shear stiffness of slabs between interior and exterior frames as a diaphragm effect may

be effective: Analysis 1.

2) Axial deformations of Bl-columns would be restrained by X-braces installed in the transverse direction; Analysis 2.

3) The stiffness of concrete should be taken

into account to calculate the "negative"

bending stiffness of composite girders in the Elastic test; Analysis 3.

4) Rigid zones in girder-to-column connection

panels must be considered, which shorten

the effective length of columns. Also

329
shear deformations of columns may be considered; Analysis 4. Each analysis, herein, was done using the results from the preceding analyses. Fig. 6 shows how the analytical results changed from Analysis 0 to Analysis 4. From this figure, it

is shown that the 3-D effects and rigid zones estimated for columns are significant to evaluate the actual story stiffness.

6. STORY SHEAR FORCE CARRIED BY BRACES AND BY MOMENT FRAMES

The ratio of story shear forces carried by

seismic resistant braces and by moment frames

is a basic parameter to design a steel structure with such braces and to evaluate the actual seismic performance, in particular,

after buckling of the braces. The ratios were calculated for each story. Here, properties of

the braces at each story are listed in Table 4. Effective slenderness ratio of the braces varied from 51.7 to 78.5 assuming an effective

length coefficient K-0.7.

Fig. 7 gives time history ratios of story shear

forces carried by the braces and by the moment

frames to those induced by the actuator forces respectively, for the 2nd story in the final

test (solid lines). This figure shows that the line on the braces turns down, and that on the moment frames turns up, having an intersection

at the response time of 7.25 sec.. About the same time the north side brace began to develop

severe buckling with the compressive strength

much decreased. This showed that the deterio- ration of the brace capacity forced the moment

frames into inelastic activity. On the contrary in the Elastic test and in the moderate test, the ratios were stable. Table 5 gives the ratios of the braces (in average) for

the two tests. These values were close to those expected in the design process. For the final test, the ratios of the braces were well predicted by the analysis shown in Fig. 7 (dashed lines). However, this prediction was considerably disturbed by local failure such as local buckling, kinks, and cracking of the braces

,

7. LOCAL BEHAVIOR IN THE FINAL TEST

7 . 1 Behavior of Braces A pair of K-braces for each story carried about half-of - the-story shear force even after the buckling of compression brace. Therefore, the overall behavior of the test building depended

on the behavior of the braces as shown in the preceding section in the final test. The axial force vs. axial displacement

relationships for a pair of the braces in the 2nd story are shown in Fig. 8. In this figure,

the numerals in the boxes, such as [1] and [2],

show the locations of the members and/or

sections in the test building as shown in Fig.

1. As seen from the figure, both braces showed

rather large inelastic deformation under the axial thrust after buckling but they did not

show tensile yielding.

The buckling strength of each brace of the test building is plotted and compared with European Column Curve (Ref. 4) in Fig. 9, where Column

Strength Curve 2 for square tubing sections with residual stresses is shown. Herein, the nondimens ional slenderness ratio X for each brace is estimated that the effective buckling

length is 70% of the clear length of each brace

(K-0.7). The test results agree well with the Column Strength Curve 2. The reason for no tensile yielding in the 2nd story braces (Fig. 8) resulted from the girder

yield at the intersection of the braces under

the vertical force produced by the difference

of the axial force of the braces, and therefore

the Increase of the tensile force of one brace

was interrupted by the reduction of the compression force of the other brace by

buckling. Fig. 10 shows the time histories of

the axial force of the 2nd story braces. The vertical displacement at the brace junction at

the 2nd floor and of the story displacement in the 2nd story of the test building is shown. These time histories shows that the increase of

the vertical displacement at the brace junction

occurred when the initiation of the decrease of

the compressive strength of the buckled brace

started, it continued until the reverse of the increase of the story displacement.

The unbalance of the axial force of the braces affected the flexural yielding of the girder

and the shear yielding of the girder panel

-

zone. The girder panel-zone showed large shear deformation; the hysteresis loop is stable as shown in Fig. 11. The maximum shear deformation amplitude reached 0.05 rad. but no cracks

were observed in and around the shear panel.

7 . 2 Behavior of Girder- to-Column Panels (Joint Panels)

The strength of girder- to-column panels (joint panels) were designed to be much less than that

of the adjacent members in this test building.

The interior joint panels in the frames parallel to the loading direction were espe- cially weaker in strength than exterior ones, relative to the strength of the adjacent

members

.

The shear deformation vs. panel moment

relationships of the interior joint panels at

the 4th floor are shown in Fig. 12. The panel

moment was estimated as the sum of the face moment of the upper and lower columns (the moment diagram was evaluated from the readings

of the strain gauges attached to the

330
intermediate sections of the colvunns) . The calculated yield panel moment according to the AIJ (Architectural Institute of Japan) steel design standard (Ref. 1) is shown by the broken

line in the figure. As seen in the figure, the joint panels show large inelastic shear deformation and the maximum amplitude of the shear deformation reached about 0.02 rad.. The plastic shear deformation in the joint panel of

the braced-bay column (joint panel [5]) concentrates on one side, because the axial thrust in

the coltunn affected the increase of plastic

shear deformation.

7 . 3 Behavior of Columns

Axial forces of the braced-bay columns became

very large when the structure was subjected to large horizontal force. Axial force vs. axial deformation relationships are shown in Figs. 13 and 14 for 2nd and 3rd story interior bracedbay columns (B2-columns) . They showed likely behavior of yielding. However, the calculated

simple axial yield strength (Ny) is about 2 times as high as the force level corresponding

to the apparent yielding.

The moment vs. axial force diagrams of both

ends of each columns are shovm in Figs. 15 and

16, respectively. Fig. 15 shows that the bottom end of the 2nd story B2 -column should

have yielded under the combination of flexural

moment and axial force. As for 3rd story

column, the combination of flexural moment and

axial force is smaller than the yield criterion

shown in Fig. 16. The joint panel adjacent to the top of the 3rd story column (joint panel

[5]) showed a large amount of yielding (Fig.

12). According to the investigation by Nakao

et al. (Ref. 6), the member adjacent to the yielded joint panel yields under a relatively

small amount of flexural moment because of the stress distribution distorted as schematically

shown in Fig. 17. This effect initiated the yielding of the top of the 3rd story column

(Figs. 14 and 16). 7.4 Behavior of Girders

All the girders were designed to act compositely with the concrete slab using shear

studs; their strength is much higher than the joint panels. Therefore the remarkable yielding was not observed in the girder. Here presented are the moment vs. cuirvature rela- tionships measured at two sections of the 2nd

floor girder in Frame-A (Fig. 18). Positive

and negative stiffnesses calculated according

to the AIJ recommendations (Ref. 2) and the stiffness of the bare steel girder are shown in

the figure. Test results show fairly good

agreement with calculated stiffnesses.

This figure shows apparent yielding under the positive bending and stiffness reduction under

the negative bending at section [8]. The

theoretical full plastic moment of the composite girder is far larger. Slippage between the steel girder and the concrete slab explains the apparent yielding in the moment

vs. curvature relationship. The composite

girder is not able to sustain the increase of

the flexural moment when the slippage occurred;

then the moment vs. curvature relationship

shows apparent yielding.

As for the stiffness reduction under the negative bending, the stiffness after the reduction is well agree with the stiffness estimated for bare steel (Fig. 18). The slippage between the steel girder and concrete

slab or between the concrete slab and the reinforcing bars may be the reason for the stiffness reduction.

The stiffness at location [9] in the figure increases in the positive bending. The reason

of this phenomenon is not yet explained.

7 .

5

Behavior of Transverse Braces

To restrain the torsional deformation of the test building under horizontal loading, transverse braces made of angle sections were

in.stalled in the exterior Frame -1 and -3 perpendicular to the loading direction. These transverse braces also restrained the vertical

movements of the column ends which arose from

the flexural deformation of the braced-bay

frame

.

The axial strains measured in the transverse

braces are presented in Fig. 19. The figure

shows that the transverse braces connected to the braced-bay columns (members [10] and [11]) were subjected to considerable strains, but

those not connected to the brace -bay columns

(members [12] and [13]) were subjected to almost zero axial strains. This occurred

because the difference between the vertical

displacements of the column in braced-bay and

of the column in Frame-A or -C are relatively

large. Yet, the difference between the vertical displacements of the column not in the braced-bay and of the column in Frame-A or -C

is quite small. Moreover, the axial strains of

the transverse braces were larger in higher

stories. The flexural deformation of the braced-bay frame produced large vertical

displacements in higher stories. From the results of the strain measurement of

the transverse braces, the three dimensional

action (transverse action) should be considered

for understanding the overall behavior of the test building.

8. CONCLUSIONS

1) The overall behavior and the local behavior

of a full-scale K-braced steel structure

were obtained with large capacity of the

331
pseudo -dynamic testing technique that simulates real earthquake responses.

2) Three dimensional effects of framing and

more accurate estimations of members' stiffness resulted in close approximation

of the analytical story stiffnesses to the experimental ones. From the consideration

of the local behavior of transverse braces,

it was shown that the three dimensional action by the transverse braces was

important to understand the overall behavior

.

3) The ratio between the story shear forces carried by the K-braces and by the moment

frames were stable and predictable before

the local failure of the braces. However, after such failure of the braces, the ratios changed so largely that further

studies are needed to estimate the behavior

of such failures.

4) Obtained local behavior was considered from various viewpoints and the behavior of many

members and sections was made clear. Furthermore, data on the members and

sections obtained in the test were verified

to be reliable and useful for attaining

knowledge about the behavior of indivi.dual members in actual structures.

9. ACKNOWLEDGEMENTS

The authors wish to express their gratitude to the members of the JTCC (the Joint Technical

Coordinating Committee of the U.S. /Japan

Cooperative Research Program; Co-chairman; H. Umemura and J. Penzien) who encouraged the authors and cordially gave advice. The authors

are also thankful to the visiting researchers

at BRI from the general contractor companies,

for their excellent help in the analysis,

tests, and data processing.

10. REFERENCES

(1) Design Standard for Steel Structures, AIJ

,

Tokyo, 1970.

(2) Design Recommendations for Composite

Constructions, AIJ, 1985.

(3) Jain, A. K. , Goel, S. C. and Hanson, R.

D. , "Hysteresis Behavior of Bracing

Members and Seismic Response of Braced

Frames with Different Proportions," Report

No. UMEE 78R3, University of Michigan,

July, 1978.

(4) Johnston, B. G., "Guide to Stability

Design Criteria for Metal Structure,"

SSRC, John Wiley & Sons, New York, 1976.

(5) Kanaan, A. E. and Powell, G. H. , "General Purpose Computer Program for Inelastic Dynamic Response of Plane Structure," Report No. EERC 73-6, University of California, Berkeley, April, 1973.

(6) Nakao, M et al . , "Strength and deformation capacity of joints and members in steel structures Part 3: Girder -to -Column Joints," Quarterly COLUMN 79, Nippon Steel Corporation, January, 1981 (in Japanese).

(7) U.S. -Japan Planning Group, "Recommendations for a U.S. -Japan Cooperative Research Program Utilizing

Large-Scale Testing Facilities," Report

No. UCB/EERC- 79/26, Sept., 1979.

(8) Yamanouchi, H. , Midorikawa, M. , Nishiyama,

I. and Kato, B. , "Full-Scale Seismic Tests

on a Six-Story Concentrically K-Braced

Steel Building - U.S.- Japan Cooperative Research Program -

, " Proceedings of Special Conference of EM Division/ASCE,

Laramie, Wyoming, August 1-3, 1984.

332
TABLE 1 MEMBER SIZE OF TEST STRUCTURE

COLUMN SCHEDULE

MARK ST0RY\^ Cl C2 C3 C4 C5

W 10 X 33 W 10 X 33 U 10 X 49 W 10 X 33 W 12 X 40

6-5 H-247.1x202.2

x7. 37x11. 05

H-253. 5x254.0

x8. 64x14. 22 H-303.3x203.3

x7. 49x13. 08

U 10 x 39 H 12 X 53 U 12 X 65 U 10 X 60 W 12 X 72 4-3 H-252.0x202.a

x8. 00x13. 46 H-306.3x253.9x8. 76x14.60 H-307.8x304.8

x9. 91x15. 37 H-259. 6x256.0

xlO. 67x17. 27 H-311.2x305.8

xIO. 92x17. 02

W 12 X 50 W 12 X 65 W 12 X 79 U 12 X 79 U 12 X 106

2 H- 309. 6x205.

2

x9. 40x16. 26 H-314. 5x306.8xll. 94x18. 67 H-327.4x310.4

xl5. 49x25. 15

W 12 X 65 W 12 X 37 W 12 X 37 W 12 X 106 W 12 X 136

1 H- 31 8. 3x308.0

xl3. 08x20. 57 H-340. 6x315.0

x20. 07x31. 75

GIRDER SCHEDULE

FLOOR\^ G1 G2 G3 G4

W 16 X 31 W 16 X 31 W 18 X 35 W 21 X 50 RFL-6FL H-403.4x140.3

x6. 98x11. 18 H-403. 4x140.3x6. 98x11. 18 H-449.6xl52.4

x7. 62x10. 80 H-529.1xl65.9

x9. 65x13. 59

5FL W 16 x 31 W 18 X 35 U 18 X 35 U 21 X 50

4FL U 18 X 35 W 18 X 35 U 18 X 35 W 21 X 50

W 18 x 35 W 18 X 40 W 18 X 35 W 21 X 50 3FL H-454.7xl52.8

x8. 00x13. 34

2FL W 18 X 40 W 18 X 40 U 18 X 35 W 21 X 50

BRACE SCHEDULE

MARK ST0Rr\^^ BR1

6 Tube 4x4x1/5.56

Box-101. 6x101. 6x4. 57x4. 57

5 Tube 5x5x1/5.56

Box-127. 0x127. 0x4. 57x4.57

4 Tube 5x5xl/4

Box-127. Ox. 127. 0x6. 35x6. 35

3-2 Tube 6x6x1 /4 Box-152. 4x152. 4x6.35x6.35

1 Tube 6x6xl/2

Box-152. 4x152. 4x12. 7x12.

7

333
TABLE 2 TEST SEQUENCE TABLE 3 , STIFFNESS OF EACH STORY

1. Free and Forced Vibration Test (VT) #1

2. Each Floor Level Loading Test (FLL) VI

3. Trial Elastic Pseudo-Oynamic Test (PSO) VI 4. Elastic Pseudo-Oynamic Test V1&2 (Maximum Input Acceleration 65 gal)

5. Pseudo-Oynamic Free Vibration Test VI

6. Trial Elastic PSD Test V2

7. Elastic PSD Test V3 (Maximum Input Acceleration 65 oal ) 8. Inelastic PSD Moderate Test (Maximum Input Acceleration 250 gal ) 9. PSO Free Vibration (PSO-F) Test V2

10. Repair Works of Girder Panel

11. Each Floor Level Loading (FLL) Test V2

12. PSO Free Vibration (PSD-F) Test V3

13. Inelastic PSD Final Test (Maximum Input Acceleration 500 oal ) lA. Each Floor Level Loading (FLL) Test V3 15. Free and Forced Vibration (VT) Test V2

Story Stiffness

6 64

5 92.5

4 127

3 152

2 133

1 220 (tonf/cm)

TABLE 4 BRACE PROPERTIES

N. Item Story\.

A

(cm^)

r

(cm)

L

(cm)

L/r KL/r

(K=0.7)

Pyn

(ton)

Pyp

(A-F)

(ton) 17.21 3.94 442.3 112.3 78.6 40.6 74.5 21.35 4.98 441.0 38.6 62.0 59.5 34.8 29.61 4.38 435.7 89.3 62.5 81.1 117.0 36.06 5.92 434.0 73.3 51.3 109.5

113.3

140.3 147.3 36.06 5.92 432.9 73.1 51.2 109.6 111.0

140.3 142.8 66.34 5.61 513.4

512.4 91.591.3

64.1 63.9192.1

191.3 292.3290.1

TABLE 5 SHARE RATIO OF BRACES IN STORY SHEAR FORCES

STORY ELASTIC MODERATE FINAL ELASTIC

ANALYSIS

DESIGN PROCESS ELASTIC ULTIMATE

6 0.59 0.59 0.59 0.67 0.74 0.76

5 0.67 0.67 0.67 0.72 0.73 0.49

4 0.71 0.71 0.71 0.73 0. 78 0.51

3 0.75 0.75 0.75-0. 51 0.76 0.32 0.56

2 0.72 0.72-0.67 0.72-0.26 0.72 0.78 0.48

1 0.83 0.33-0.65 0.83-0.64 0.81 0.35 0. 58
LOADING DIRECTION

Ooan

K-brac£ ooon

1 2 3

I 7500 I 7500 I

MU

*—[6] /\

XX

1 2 3 frame B ABCframe 1

1 2 3 frame A, C A B C frame 3

FIG. 1 PLAN AND ELEVATION OF TEST STRUCTURE

335
0 . 0 S. 0 10-0 15.0 TIME ISECI

20. 0

FIG. 2 FLOOR LEVEL DISPLACEMENTS (MODERATE)

336
OISPL

iCm

25 . 0 RF

15 . 0 TIME ISECI

0.0 5.0 10.0 15.0

TIME ISECI

IIME ISECI

FIG. 3 FLOOR LEVEL DISPLACEMENTS (FINAL)

337
0I5PL icm

- 1.00 +

A- A Analysis 2 Analysis 3

•

• Analysis 4

FIG. 5 HYSTERESIS MODEL OF BRACES FIG. 6 DEVIATIONS OF ANALYTICAL STIFFNESS FROM THE TEST tJFlSU

338
2 STORY

FIG. 7 RATIOS OF STORY SHEAR FORCES CARRIED BY BRACES AND BY MOMENT

FRAMES TO THOSE INDUCED BY ACTUATOR FORCES

FIG. 8 AXIAL FORCE vs. AXIAL DISPLACEMENT

RELATIONSHIPS OF K-BRACES IN THE 2ND STORY

339
o -152.4x152.4x12.7 • 0-152.4x152.4x6.35 A D-127.0 x127.0x 6.35

X -127.0 X 127. Ox 4.57

FIG. 9 BRACE BUCKLING STRENGTH

FIG. 10 TIME HISTORIES OF BRACE AXIAL FORCE, VERTICAL DISP. AT BRACE JUNCTION AND STORY DISPLACEMENT

340
Q (l-on) tIOO

•0.1 -0.05

Qy = 75.8

100 FIG. 11 SHEAR DEFORMATION OF THE GIRDER

PANEL-ZONE AT THE 3RD FLOOR

FIG. 12 SHEAR DEFORMATION vs. PANEL MOMENT RELATIONSHIPS

OF THE GIRDER-TO-COLUMN PANELS AT THE 4TH FLOOR

N comp (ton)

FIG. 13 AXIAL FORCE vs. AXIAL DEFORMATION RELATIONSHIP

(2ND STORY 82 -COLUMN)

N comp (ton)

FIG. 14 AXIAL FORCE vs. AXIAL

DEFORMATION RELATIONSHIP

(3RD STORY 82 -COLUMN)

341
Ncomp/Ny Ncomp/Ny

Ncomp/Ny

FIG. 15 M-N DIAGRAM

(2ND STORY B2-COLUMN)

Ncomp/ Ny

3RD STORY B2 COLUMN [T] FIG. 16 M-N DIAGRAM

(3RD STORY B2 -COLUMN)

Q-Q section

FIG. 17 DISTORTED STRESS DISTRIBUTION

342
Mp(neg)= 48 .3 t-m

FIG. 18 MOMENT vs. CURVATURE RELATIONSHIPS OF THE 2ND STORY GIRDER

base shear

(ton)

mic

250 base shear

( ton) -250

250 -250

micro strain -250

0 250

micro strain -250

FIG. 19 STORY SHEAR FORCE vs. STRAINS IN THE TRANSVERSE

BRACES RELATIONSHIPS

343
FULL-SCALE STUDIES OF STEEL STRUCTURES

by Robert D. Hanson^

ABSTRACT

The us/Japan Cooperative Research

Program in Earthquake Engineering

Utilizing Large Scale Test Facilities

was initiated in 1980 under the auspices

of the UJNR Panel on Wind and Seismic

Effects. The first phase of this program focused on a reinforced concrete

test specimen, the second phase on a structural steel test specimen, and current program on masonry specimens. A summary of the USA activities and results in the Structural Steel phase of the program is given here. This

information is extracted from the papers presented at the Sixth Joint Technical

Coordinating Committee Meeting held in Maui, Hawaii from June 20 - 25, 1985 and does not include direct participation by

Douglas Foutch, Subhash Goel and Charles

Roeder in the full scale specimen tests

in Japan. At this meeting nine major

accomplishments of the program were identified and nine recommendations for future action were made. Because many

of these actions require additional

funding from the sponsoring government

agencies and support of the UJNR Panel

on Wind and Seismic Effects they are included herein in their entirety.

1. INTRODUCTION

Shortly after the US/Japan Cooperative

Program was established with funding

from the USA National Science

Foundation, and the Government of Japan

Ministry of Construction (Building Research Institute) and Science and Technology Agency (Disaster Prevention

Research Institute), a Joint Technical

Coordinating Committee (JTCC) was formed

to provide and encourage cooperative

technical interaction of all research investigators and to provide independent

review of key technical decisions in research coordination. Professor Hajime

Umemura and Professor Joseph Penzien

have served as JTCC Chairmen and Professor Makoto Watabe and Professor

Robert D. Hanson have served as JTCC Technical Coordinators for both the reinforced concrete and structural steel phases of the program. Other members of

the JTCC were principal investigators of the individual research tasks in Japan

and the United States and prominent

technical leaders. The independent members for the USA were Henry J. Degenkolb, Paul C. Jennings, Roy G. Johnston and Roland L. Sharpe. The activities and milestones of the Cooperative Research Project are reflected in the papers and discussions

held at the JTCC meetings. The First JTCC meeting was held in Tsukuba during

January, 1981 prior to testing of the Full-Scale seven story reinforced concrete specimen. From this meeting

the testing program for the specimen was agreed to be a single-degree

pseudo-dynamic test program at four increasing levels of displacement

response. The Second JTCC meeting was

held in Tsukuba during July, 1981 to discuss the results of these four tests

and to plan- the repair and retests, nonst ructural component additions, and preliminary designs for the steel test specimen. The Third JTCC meeting was held in Tsukuba during July 1982 to complete the design of the full-scale structural steel specimen, to review associated reinforced concrete and steel tests and analyses, and to develop a complete program of correlation studies

for the reinforced concrete program.

The Fourth JTCC meeting was held in Tsukuba during June, 1983 to review the construction of the steel specimen, to establish the loading program for the full-scale specimen for multiple-degree

pseudo-dynamic testing, and to review associated support tests and analyses of the reinforced concrete and steel studies. The Fifth JTCC meeting was held in Tsukuba during February, 1984 at the midpoint of the eccentric braced

specimen test program. The earlier

concentric braced specimen and eccentric

braced specimen test results were presented, a program to complete the eccentric braced specimen tests was

^Dept. of Civil Engineering, Univ. of Michigan , Ann Arbor, MI 48109-2125

344
established, the nonst ructural test program was reviewed, and associated

supporting research activities were

evaluated. The Sixth JTCC meeting was

held in Maui, Hawaii during June, 1985

to review all support tests in the USA

and Japan for the reinforced concrete

and structural steel programs, to review

the test results from the full-scale

steel specimen including the nonstructural component phase, to discuss analytical studies, and to plan

the activities necessary to complete

these phases of the Program. Because the preceding paper in this meeting provides data on the full-scale

structural steel specimen tests, the following will provide only a brief

summary of the USA associated support

activities as reported to the JTCC at the Hawaii meeting. As a conclusion the 6JTCC Summary, Resolution,

Recommendations and Accomplishments is quoted in its entirety.

2. EARTHQUAKE SIMULATION (SHAKING

TABLE) OF 0.3 SCALE MODELS OF FULL-SCALj: SPEC IMEN

A dual structural system, consisting of

a moment-resisting space frame and braced frame, was selected for the Test Structure of this phase of the program.

Two different types of braced frames

were studied; the first was based on the use of a concentric K bracing

system; and the second used eccentric

braces. The studies that were conducted

at the University of California,

Berkeley under the direction of Vitelmo

Bertero can be grouped as follows: (1) Preliminary analytical studies conducted

to review the soundness of the preliminary design of the building and

to produce the analytical information

required to determine the largest scale

model for the two test structures that could be tested on the shaking table at Berkeley; (2) The final selection of the largest scale model (selected to be 0.305) and its design, the selection of the steel and of the concrete to be used

and the problems encountered in the fabrication of the different components

of the model, and in their final assemblage and in their instrumentation;

(3) The experimental and analytical

studies conducted regarding the simulation and behavior of the composite

construction used for the floor system

of the two test structures; (4) Determination of the mechanical

characteristics of the 0.3 models during

the different stages of its construction; (5) Experiments conducted

on the 0.3 scale models at the Berkeley ' earthquake simulator facility; (6) Comparison of experimental results with

those obtained for the full-scale model

tests conducted at Tsukuba, Japan, and those predicted analytically; and (7) Analysis of the implications of the results obtained on the states of the

art and of the practice of seismic

resistant design and construction of steel building structural systems based

on use of braced and unbraced moment

resistant frames. Conclusions and correlations of the 0.305-scale dynamic

test results with the full-scale pseudo

dynamic tests remain to be done.

3 . CYCLIC TESTS OF FULL-SCALE COMPOSITE

Cyclic tests of full-scale beam-column

assemblages with composite metal deck

and concrete slab was performed at> Lehigh University under the direction of Professor Le-Wu Lu. Both interior and exterior joints with beams framing into

the strong and weak directions of the columns were tested. These subassemblage test results will be compared and correlated with similar

joints from the full-scale specimen

tests in Tsukuba.

4. CYCLIC BEHAVIOR OF COMPOSITE FLOOR

SYSTEMS IN ECCENTRICALLY BRACED FRAMES Eccentrically braced steel frames can provide high elastic stiffness and large

inelastic energy dissipation capacity. This makes this type of structural system a viable alternative for seismic

design. Performance of these type of systems depends to a great extent on the behavior of the link sections in the

345
beams. The behavior of bare steel links

has been extensively investigated in the

past, however, the effects of the composite floor system in eccentrically

braced frames has not been addressed.

An experimental investigation on bare steel floor beams and composite floor beams in eccentrically braced frames were made at the University of California, Berkeley under the direction

of Professor Egor Popov. Behavior of both types of floor beams were subjected

to cyclic loading with emphasis on the performance of the link. The performance of a composite floor beam's

response to a simulation of the Taft earthquake was also made. The test results for the bare steel and composite

floor beams were evaluated by comparing

the elastic stiffness, strength, and ductility of the links, in addition to the restraint of the link by the slab. The effective width of the composite

floor beams under cyclic loading was determined by measuring strain profiles.

The main conclusions were: (1) Composite

short links designed as bare steel short

links yield in shear and dissipate

energy primarily through web yielding;

(2) Composite short links have a greater

yield and maximum strength compared to bare steel links under cyclic loading;

(3) The composite floor system in eccentric braced frames increases the elastic stiffness of short links, in addition to effecting the rotational

stiffness of the floor beams which frame

into the ends of the link; (4) The composite floor system is not effective

in providing restraint at the ends of the link to prevent lateral torsional

buckling of the steel section. Transverse beams should be used to brace

both ends of the link; (5) The shear lag effect is present in the composite floor system under cyclic loading. The longitudinal strain in the slab has a wider profile for positive moment

compared to negative moment. The width

of the profiles increases in slab sections through the link, whereas

outside the link the profiles remain a constant width as the amplitude of link

distortion is increased; (6) The amount

of link deformation to cause web buckling is independent of the link's deformation history, hence it is not a function of the amount of energy dissipated by the link. Web buckling

occurs when the link deformation exceeds

a specific accumulated value of deformation beyond the last zero shear crossing within the cycle of buckling.

The specific value is related to section

dimensions and stiffener spacing; (7) The web stiffeners should be designed in order to satisfy both the strength and rigidity criteria in order to avoid unfavorable post buckling performance of the link; and (8) The maximum forces developed in the jacks testing composite

short links were greater compared to jacks forces developed in bare steel short link specimens. The forces in composite link specimens were, on the average, greater in V-brace EBF simulations compared to K-brace EBF simulations. To assure that braces

remain elastic and do not buckle, the brace forces from a bare steel EBF collapse mechanism should be increased

by a factor of 1.6 to 1.8 when designing

the braces.

5. SMALL-SCALE MODEL TESTS OF STRUCTURAL STEEL ASSEMBLIES

In the steel phase of the project, tests were conducted on 1:12.5 scale specimens

of beam-column assemblies, a braced

frame, and an unbraced frame at Stanford University under the supervision of Professor Helmut Krawinkler. All of the small-scale specimens included floor slabs which consisted of concrete

reinforced with wire mesh and were cast onto metal deck. Shear studs welded to the beams were used to make these slabs

work compositely with the beams. The beam-column assemblies represent

portions of the frame on the center column line and of the exterior frames. The braced frame specimen represents a model of the center unit of the concentrically braced prototype test structure including the floor slabs. The unbraced frame specimen is the braced frame specimen with the braces

346
removed. This specimen was tested to give information about the resistance of the moment frame in the braced frame

specimen.

The major goal of this study was to assess the feasibility and limitations

of small-scale model testing in earthquake engineering by testing models

of steel structural assemblies and correlating the results with prototype

tests. This correlation was made

possible by the U.S. -Japan Cooperative

Research Program in Earthquake

Engineering in which full-scale prototype structures and components were

tested. All small-scale models tested in this study were constructed at a scale

of 1:12.5. The three beam-column assemblies were

replicas of full-size assemblies tested

at Lehigh University. The prototype test

of an exterior assembly with a column

flange connection resulted in severe

inelastic column panel zone deformation

and ultimately fracture of the beam bottom flange next to the joint. This behavior was reproduced in the model

tests. In the prototype test of an exterior assembly with a column web connection little inelastic shear deformation occurred in the joint region

and the specimen deformed by inelastic

flexure of the beam' and column. Failure

was by fracture of the lower beam

flange. This was also reproduced in the model test. The prototype test on an interior

assembly with column flange connections

resulted in severe inelastic shear distortion of the panel zone and ultimately fracture of the bottom beam

flanges caused by high localized

curvatures at the corners of the severely distorted joint. In the model

test the inelastic deformations in the panel zone were reproduced, but the column web panel tore before the beam

bottom flanges had flexed sufficiently

to fracture. The correlation between the results of the model braced frame test and the

prototype test was limited by the compromises which had to be made in loading and also by the absence of the two exterior moment frames in the model

specimen. Despite these compromises the damage pattern in the prototype

structure, represented by brace

buckling, beam hinging, joint

distortions, and ultimately fracture of the tubular braces, was reproduced in the model test. The model test predicted larger peak loads than the prototype test. This discrepancy is mostly due to the excessive strength of the model bracing members.

Based on the results of this study, the following major conclusions on small-scale modeling and structural

behavior were drawn: (1) The experiments

indicate that the overall cyclic

inelastic load-deformation response of components and structures can be reproduced adequately in tests of carefully designed and detailed

small-scale models; (2) The steel beam-column tests have shown that inelastic panel zone distortion and beam

flexure can be properly simulated at small scales. In addition, the action

of the composite floor slab can be simulated as shown by the load -deflection and moment-strain diagrams;

(3) The braced frame test has demonstrated that the post-buckling

response of complex bracing systems can

be reproduced in model tests. However,

this test has also shown that the buckling strength of braces is sensitive

to material properties and connection

details, which are parameters that are difficult to simulate accurately at small model scales; (4) Size effects in small-scale model tests will usually

result in high predictions of strength

(because of strain gradient effects) and ductility (because of improved fracture

properties) under cyclic inelastic

loading. In particular the improved

fracture properties, together with

fabrication problems, render model test results questionable for an evaluation

of localized failure modes in structural

connections; (5) and The small-scale

model tests led to the same major

347
conclusions on structural behavior as the prototype tests. Amongst these conclusions the following deserve

emphasis: (5a) Panel zone shear strength

may limit the capacity of steel beam-column joints. Inelastic shear deformations of the panel zone are usually very ductile, absorbing

considerable energy before failure

occurs. Excessive shear distortions in joint panel zones may cause fracture of the beam flange welds due to the high curvatures at the joint corners; (5b) Composite floor slabs increase the strength of beams and joints in steel

frames. At beam-column joints this increase in strength is limited by the bearing capacity of the concrete that transfers the slab stresses to the column. Under severe deformations the composite floor action is seen to deteriorate as the concrete surrounding

the column is crushed; (5c) The strength

of "K" braces in a braced frame deteriorates rapidly after first buckling. This will cause the formation

of a "soft" story which may then be subjected to excessive drifts; (5d) Tubular braces are susceptible to severe

local buckling after brace buckling.

Under load reversals this local buckling

may lead to crack initiation and fracture of braces.

6, PSEUDO-DYNAMIC TESTING: CONTROL

ISSUES AND PERFORMANCE EVALUATION

The pseudo dynamic test method provides

a means for testing of full scale structures subjected to simulated

seismic ground motions. The method uses

an on-line computer to calculate the appropriate displacement response of the structure, based on an analytical model

of the inertia forces and the reaction

forces due to the seismic ground

acceleration and on direct measurements

of the inelastic restoring forces. The structure is quasi-statically caused to deform according to the computed

displacment response by use of electrohydraulic actuators. Experimental

and theoretical research at The University of California, Berkeley under

the direction of Stephen Mahin and at

The University of Michigan under the direction of Professor Robert D. Hanson

have been cooperatively exchanged with

colleagues at The University of Tokyo

and The Building Research Institute

(Tsukuba)

.

The pseudo-dynamic test method has been demonstrated to be a powerful and versatile procedure for assessing the seismic performance. In particular, it may provide insight on the seismic

response of structures, that are too

large, massive or strong for testing

using available shaking tables. In addition, extension of the basic

procedures will permit tests considering

complex support excitations not possible

with other means. However, like all experimental

procedures, the pseudo-dynamic method

has inherent limitations and errors. These must be fully recognized in developing pseudo-dynamic test programs.

In particular, the need for high

performance control equipment and instrumentation must be taken into account.

A variety of future developments and investigations should be undertaken to make the pseudo-dynamic method more convenient and reliable. For example,

the various implementation software algorithms developed in the U.S. and in Japan should be refined and integrated

into a convenient package available to those interested in pseudo-dynamic

testing. Various routines for self diagnosing of experimental errors, analytical substructuring, computer

graphics, data analysis and etc. should

be included. Guidelines for the use of such algorithm should be established.

Additional research is needed related to numerical methods. This is especially

true for methods associated with

controlling the adverse effects of experimental errors and with

substructuring methodologies.

Experimental verification of these methods is essential.

One of the most important aspects to be

348
researched relates to understanding and improving the ability to control test specimens. At present, there are very

few reliable guidelines available for selecting appropriate experimental

equipment, or for establishing the various controller parameter settings, needed to assure test stability and accuracy. New controller designs to insure stability and to mitigate cross coupling effects in stiff systems should

be investigated. In particular, digital controllers capable of combining the inner and outer control loops and implementing hybrid analysis

capabilities in "real time" should be assessed. Not until reliable guidelines

for the use of pseudo-dynamic test equipment are established can the method

be used dependably.

For most structures tested by the pseudo-dynamic method, the response is dominated by the fundamental frequency.

However, spurious higher-mode responses

can be excited by experimental errors. While high performance experimental

equipment can be used to reduce these

errors, their effects can be severely

amplified if relatively large

higher-mode frequencies are present in the test specimen. Under such condition,

viscous or numerical damping is desired

for eliminating the spurious higher-mode

responses.

It is desired that the viscous damping

ratios selected for the higher modes be much larger than that for the fundamental mode. Furthermore, since the nonlinear stiffness properties of that specimen are not accurately known, initial stiffness-dependent viscous

damping has to be selected. Based on this approach, the damping

characteristics may change as nonlinear

deformation occur. However, it is imperative that these characteristics

are not undesirably altered. To evaluate

the variation of effective modal dampings in a nonlinear system, the complex eigen-value and approximate

modal analysis techniques have been used

to uncouple the equations of motion.

The results of these studies show that

the higher-mode dampings do affect the response of the fundamental mode during

nonlinear oscillations. In fact, for a softening-type non-linear system, the effective damping on the fundamental

mode can be greatly increased by

selecting large higher-mode dampings.

In view of the above problem, a modified

Newmark's integration algorithm with

frequency-proportional numerical damping

has been proposed. It has been shown

that the accuracy of this algorithm is comparable to that of the explicit Newmark's method. For softening-type

nonlinear systems, a criterion has also

been developed to ensure that numerical

energy dissipation always remains positive. This preserves numerical

stability as the fundamental frequency

of a specimen is reduced by nonlinear

deformations. Furthermore, based on the energy criterion developed, the numerical damping on the fundamental

mode can be kept zero or arbitrarily

small during nonlinear oscillations.

This method, thus, seems to be more attractive than the initial stiffness-dependent viscous damping.

7. ANALYTICAL MODELS FOR FULL-SCALE

SPECIMEN

A six-story, two bay, full-scale steel structure with eccentric bracing was tested by pseudo- dynamic method at the Building Research Institute, Tsukuba,

Japan, during January-March, 1984. The

two exterior frames are unbraced moment-resisting frames with one column

in each oriented for weak-axis bending,

and the interior frame is braced with

eccentric K-bracing in the north bay. The floor system consists of formed metal decking and cast-in-place

lightweight concrete which acts compositely with the girders and floor

beams. The main pseudo-dynamic testing of the structure consisted of three parts. S21°W component of the 1952 Taft earthquake was used in the elastic and inelastic tests with peak acceleration

of 65 and 500 gals, respectively, (1 gal

349
= 1 cm/sec^). The final large amplitude

test was performed as steady state responses to three sinusoidal ground

acceleration pulses of one cycle each. Based on the analytical study of the elastic and inelastic response of the Phase 2 structure at The University of Michigan under the direction of Professor Subhash Goel/ the following

conclusions can be drawn: (1) The refined analytical model seems to represent the global behavi’or of the structure quite accurately. The agreement between the analytical results

and those from the elastic and the inelastic tests was excellent; (2) The elastic response seems to be very sensitive to the fundamental period; and

(3) The assumed characteristics of the hysteresis behavior of shear links had significant influence on the inelastic

response. More information is needed on the behavior of these elements.

8. SYSTEM EVALUATION OF ELASTIC PSEUDO-DYNAMIC FULL-SCALE TEST DATA

The low-amplitude pseudo-dynamic test data from a six-story, two bay full-scale steel structure with

eccentric bracing was analyzed using a system identification technique at the California Institute of Technology under

the direction of Professor James L. Beck. Properties of the first three modes of the structure were identified

and the calculated response due to these modes when subjected to the earthquake

ground accelerations used in the test

was found to be in excellent agreement

with the displacements and pseudo-velocities from the test. However, the pseudo-accelerations from

the test did not agree well with those calculated from the identified three-mode model because the third mode

gave unreasonable large accelerations in the test. This is attributed to an instability in the experimental method

that is produced by the cumulative

effects of feedback of errors in the computer control system. The identified

natural periods of the first three modes

of the structure were 0.553, 0.191 and

0.102 second. The following conclusions were made from

this study: (1) The feedback of control

and measurement errors into the computer

model used to produce the experimental

displacements in the elastic test produced a cumulative error of large

amplitude in the pseudo-acceleration at frequencies around 10 Hz. However, these errors have an insignificant

effect on the displacements and pseudo-velocities, where the high-frequency content was greatly

reduced; (2) For similar amplitudes to those in the elastic test (peak

acceleration and displacement of 30%g and 1.4 cm respectively at the roof), a linear model based on two classical

normal modes should give an excellent approximation to the response of the structure; and (3) At these amplitude

levels, the equivalent viscous damping

factors for the first three modes are

1.2%, 2.1% and 2.2% of critical. The damping of the fundamental mode is consistent with the value of 1.23% reported for a pseudo-dynamic free vibration test performed after the ®l3Stic test. Part of the damping was artificially introduced as numerical

viscous damping in the computer model

used to produce the experimental

displacements. Also, the displacement

overshoots programmed into the computer

control system apparently contributed

some damping. The energy dissipated by the structure itself produced equivalent

viscous damping factors less than 0.8%,

1.7% and 0.2% for the first three modes. These low values of damping compared

with values of 3% to 8% from the earthquake response of tall buildings in the field is most likely due to the absence of energy dissipation by nonstructural components and the smaller

amount of radiation damping produced by

the relatively stiff foundation for the test structure.

350
9. SUMMARY, RESOLUTION, RECOMMENDATIONS

AND ACCOMPLISHMENTS - 6JTCC

9.1 Summary

The Sixth Joint Technical Coordinating

Committee Meeting for U.S. -Japan

Cooperative Research Program Utilizing

Large-Scale Testing Facilities was held

at Kaanapali, Maui, Hawaii, U.S. A., from

June 20th through June 25th 1985. Thirteen U.S. and five Japanese JTCC members plus twelve observers participated in the meeting.

On the first day of the meeting (June

20th), after the opening session, (1) support tests in both countries and full-scale tests for RC phase were all reported, and then (2) future tasks for

RC phase were discussed. On the second

day, (1) steel phase on full-scale tests

of concentric K brace system:, (2) of eccentric K brace, as well as (3) of moment frame system, and then (4) U.S. steel support test results were reported

and discussed. On the third day, (1) the latter half of U.S. steel support test results and (2) behaviors and analytical

results on non-structural elements of steel full-scale tests were reported and discussed. On the fifth day (June 24th

Monday) , Japanese steel support test

results, as well as (2) analytical

studies on steel full-scale tests were

reported and discussed. Brief

introduction of recent aspects on masonry phase was reported.

Following the discussion on future tasks

of steel phase, accomplishments of the program were identified and potential

future tasks of U.S. -Japan earthquake

engineering cooperative research program

were discussed. On the final day, at concluding session, the statement of accomplishments was agreed and the following resolution and recommendations

were adopted.

9.2 Resolution

Members of the JTCC recognize this meeting to be the tenth anniversary of

the Review Meeting of the U.S. -Japan

Cooperative Research in Earthquake

Engineering with emphasis on the Safety

of School Buildings, also held in the State of Hawaii in 1975, at which

meeting the recommendation for the current U.S. -Japan Cooperative Program

in Earthquake Engineering Utilizing

Large-Scale Testing Facilities was compiled and signed, marking the very beginning of the project. They agree

that the meeting was successful and fruitful for both countries in a cordial

atmosphere at Kaanapali, Maui, and they

are pleased to acknowledge the fine arrangements for the meeting and the great effort and excellent performance

of all principal investigators, and in particular, the continuing encouragement

and guidance by Professor J. Penzien and

H. Umemura.

(1)

The experimental and analytical

accomplishments of the program have important consequences to the practice of earthquake resistant

design and earthquake engineering

research. Reports of individual

research efforts and correlation

studies should be completed at the earliest possible time. The implications of these studies should

be prepared for the design and research professionals as soon as possible.

(2) The reports, papers, photographs and data developed in this program

should be fully documented and made available according to the schedule

given in Attachment 1.

(3) The Structural Steel program must be completed at the earliest possible

time by performing the necessary

correlation studies amongst the various experiments and with

appropriate analytical studies. These results should be reported as recommended in (1) and (2) above.

(4) Exchange of research investigators

should be implemented to hasten data

351
interpretation and correlation study

completion. These individuals who were active participants in the experimental research program will provide the most effective exchange.

(5) The results of the experimental and analytical studies in this program

have provided significant benefits

as outlined in the ACCOMPLISHMENTS.

The results and specimens of the current program can be effectively

and economically utilized to investigate additional variations of framing suggested from analytical

parameter studies. This is a cost effective extension of the existing

program.

(6) The possibility of new phase for a cooperative program should consider

systems such as (a) mixed structures

and (b) inventive bracing systems. A planning group for the new phase should be organized.

(7) For quicker and smoother information

exchange to achieve well-coordinated

reports and papers, frequent

national JTCC meetings should be held in each country.

(8) A seventh JTCC meeting is necessary

for correlation of the various

aspects of the U.S. and Japanese

support programs and to discuss the significance of preliminary

conclusions. The meeting should be held in California during the last quarter of 1986.

(9) To carry out the above recommendations, the sponsoring government

agencies in both countries are urged

to provide the necessary funds on a timely basis.

9.4 Accomplishments of the U.S. -Japan

Cooperative Research Program

U-tilizinq Large-Scale Testing

£ag.Uitieg

This unique research program combines

full-scale, small-scale, subassembleage

and component test with analytical

investigations of specific building

structures. These experiments were carried out using static, static cyclic,

on-line computer and dynamic shaking

table tests at unique facilities in the United States and Japan by the best researcher available in the two countries. The coordinated research

activities, scientific and personnel

exchanges have produced results which

could not have been achieved by

individual efforts in the U.S. of in Japan. The participants at the 6th JTCC have identified some of the major

accomplishments achieved at this time

and provide the following list arranged

by scientific and technical

accomplishments. This list is not meant

to b exhaustive and it does not contain

those anticipated from the final stages

of the structural steel phase of the program.

(1) Significant. Contribution to the

U.n(geg5tan4ing <?f ths Rg.^l B ehavi or of Buildings During Earthquake

These test structures closely resembled

real dual framed buildings. The contributions and inter- action of these

dual systems are more fully appreciated

and may lead to better building with

less cost. The incorporation of interior and exterior walls, glazing and cladding added stiffness to the structure, thereby changing the dynamic

characteristics of the building. The static tests of these elements provide

insight to the large deformation

characteristics.

(2) Demonstration of the Need for late^ated Analytig^ l and

Sxpgriingntal Studie? of Stcuctural Behavior

In the elastic range it is dangerous to rely on prediction of response on a deterministic basis because of the sensitivity of the response to building

period and the sensitivity of the building period on mathematical models

352
of the structure which may not include

adequate three dimensional effects,

shrinkage, loss of composite action, and axial load effects of dead and live

loads. In the inelastic range initial damage will predicate the subsequent

response and damage. Our current models

are not sufficiently developed for independent reliance on analytical

studies for systems not experimentally

verified.

(3) On-Line Computer (Pseudo-Dvnamicl

Xest Control Me thod. The traditional static cyclic test methods are predetermined before the test begins. They are not capable of reflecting the changes in force and displacement distributions caused by

inelastic action or damage to portion of the structure. Realistic loading

patterns are essential to accurately

represent earthquake loading and to follow the resulting inelastic response.

(4) Confirmation of the Validity and Reliability <?f Scal9-M9.«jg.ls- £.ai: Studies of Overall Structural Response

This cooperative program with full size

and various scale models have confirmed

the validity and reliability for overall structural response, and this has important economic consequences for future structural testing.

(5) Confirmation of the Validity and Reliability o£ Large-Stale CqmpQnent

Tg.s.ts. for Studies of Lotal Struetural B^haviec

The program has shown that large-scale

component tests are necessary to define

local detail behavior. In designing

these tests it is important that rational load programs should be established for that portion of the structure.

(6) Three-Di ipeneional E ffett g.af. Mopo l.i t.l:>ic... St i:u ot ui:e? Can Not

N,eg,le.qt.e.<j. Actual behavior of buildings requires

analytical and experimental modeling to realistically represent the three dimensional characteristics even for - planar loading. The necessity for this modeling for multi-directional

earthquake considerations is selfevident

.

(7) A.dy.aiLg£.s-in Strudtucal. Te?tina

Technology and Efficient Utilization

of Experimental Facilities and

The coordinated testing program, involving full-scale and reduced-scale

structural components, assemblies and complete frames and utilizing

quasistatic, pseudo-dynamic and truly

dynamic (earthquake simulators) testing

techniques has demonstrated the advantages and disadvantages of different experimental approaches. In addition, the problem has resulted in significant advancements in the technologies of testing large-scale

multi-story structures. These advancements will make possible more efficient and cost-effective

utilizations of existing and planned

experimental facilities in the future.

(8) s.yngLgifiJt.ic.-,6.££g-g-t <?£ Co<?cjin^tgd

Multi-Institutional Cooperative

Rgggacch Prbg.Laros. The benefits of active coordinated

interchange between key research personnel during the execution of a research program directed at a common

goal has been demonstrated. The utilization of the unique research facilities at specific institutions can

be effectively utilized in this manner. This provides an economical utilization

of personnel and facilities.

353
(9) The U.S. -Japan Cooperative Research Program Has Developed an Open, Full Exchange of Scientific Research

Ideas, Concepts and in Earthquake

Engineering

This program has developed a mutual

respect and trust among the U.S. and Japanese researchers. This relationship

allows the open, full exchange of technical information beyond the immediate scope of the research program.

This relationship should be maintained

to the mutual benefit of both countries.

354
THEME V

Mexico City and Chilian Earthquakes

EARTHQUAKE DAMAGE IN MEXICO CAUSED BY THE SEPTEMBER 1985 EARTHQUAKES

BY Edgar V. Leyendecker

William C. Stone

Pel ix Y. Yokel' ABSTRACT Following the September 19, 1985 Mexico

City earthquake, a team consisting of four engineers and one seismologist from the National Bureau

of Standards (NBS) and the United

States Geological Survey (USGS) was dispatched to Mexico City to provide

technical advice to the US rescue effort and to asses structural

damage. This report Is primarily based

on data gathered by the team, but it also contains a compilation of other available Information. The report addresses the origin and

characteristics of the observed ground

motion, the ability of buildings

designed in accordance with present and proposed seismic design provisions to resist this type of ground motion, and observed data on structural and

f oundat Ion fail ures.

On September 19, 1985, an earthquake of magnitude 8.1 struck the western coast

of Mexico near the town of Lazaro

Cardenas in the State of Michoacan (See

figure 1). A team of engineers and seismologists from the National Bureau

of Standards and the United States Geological Survey was sent to Mexico

following the earthquake to provide

technical assistance and to study Its effects. Despite a distance of more than 400 kilometers from the epicenter, the earthquake caused widespread damage in central Mexico and was felt as far

north as Houston, Texas. The city was shaken on Friday, September 20 by a magnitude 7.5 aftershock with an epicenter 100 kilometers southeast of the main earthquake.

2. background

The earthquakes which struck Mexico

City on September 19 and 20 originated in a narrow seismic zone that runs 2500 kilometers along the western coast of Central America from Panama to central Mexico west of Puerto Vallarta. A zone (figure 2) of Intense seismicity has been formed where the Cocos plate dips beneath

the North American plate, giving rise

to earthquakes along the coast and volcanoes inland in central Mexico. Mexico City lies in a broad basin

formed approximately 30 million years ago by faulting of an uplifted

plateau. Subsequent volcanic

activity closed this basin resulting

in the formation of historic Lake

Texcoco upon which the Aztec capital was built. The expansion of Mexico City by the Spaniards led to the gradual

draining of Lake Texcoco. With a population of about 18 million,

Mexico City is the world's largest

population center and the site of more than one million man-made

structures. The photograph in figure

3 was taken a few days after the earthquake and shows I ittle evidence

of widespread damage. In fact about 330 buildings were severely damaged

or collapsed as shown in figure 4. Buildings in the 6 to 15 story range were particularly hard hit. Some of the reasons for the damages are

discussed in this paper. Today, much of the city rests on lake deposits (figure 5) made up of highly

plastic clays and sands with

thicknesses of more than 50 meters. These lake deposits have long been identified as a factor making Mexico

City particularly vulnerable to earthquakes. When unconsol Idated material underlies a site, the

seismic shaking is modified from what

it would have been on competent rock. The general trend In the lake bed region of Mexico City is for

National Bureau of Standards, Gaithersburg, Maryland

20899

357
unconsolidated sediments to increase

the ampi itude of ground shaking at higher periods leading to what Is generally termed soil amplification.

In the center of the historic district,

where many buildings between five and

20 stories high did match the natural period of the quake, devastation was widespread. Dots shown in figure 6 mark the locations of damaged or destroyed buildings. Practically all of the most seriously affected area was confined within a five by five

kilometer area just east of the U.S. Embassy.

3. SIR.U.QIU.R/\L. P.E RE:Q.£M AMCE The twelve story reinforced concrete

structure shown in figure 7 housed the Ministry of Communications and

transport, and the nation's main

microwave transmitter. Failure of this structure precipitated a near total

col lapse of long-distance telecommunications between Mexico City and the rest of the world, a factor which

complicated the coordination of

International rescue efforts. The SCT acce 1 erograph station (figure

8), located on the grounds of the telecommunications center, provides a good indication of response in the historic district, where most of the structural damage occurred. At the communications center the predominant

acceleration component was east-west with a period of approximately two seconds. Vertical accelerations were

neg I I g

I b I e

.

The East/West ground motion records

(figure 9) indicated peak free field accelerations of 0.17g with peak to peak ground displacements of 42 cm. because of the long duraton of the Earthquake structures with a natural period close to the predominant two second period of the earthquake had ample opportunity to be driven to resonance.

The reason why so many apparently well- designed structures collapsed in the central portion of Mexico City is best explained by the design response

spectrum from the communications center site (figure 10). Structures with

natural periods in the vicinity of two seconds, which is about 5 to 20 stories, were subjected to lateral accelerations of up to one G, nearly

six times the highest lateral

acceleration previously recorded. This large amplitude, combined with the

exceptionally large number of stress reversals, contributed to the

majority of the collapses. The solid horizontal I ine in figure lOshows the current Mexico city building code

design spectra for non-ductile

frames, which was based on previous

earthquake records for the valley of Mexico. Thus, even buildings

designed entirely in accordance with

the code were severely under strength

for this quake.

Hospitals are one of the most

critical facilities in any disaster. However, major portions of three of the city's largest hospitals

collapsed. Some 1200 people were

buried as six buildings collapsed at the Centro Medico and Mexico General Hospital (figure 11), one of the largest medical complexes in Latin

Amer I ca

.

Yet another 400 medical personnel and patients were trapped in the

Maternity wing of the Juarez

Hospital, eight blocks south of the Presidential Palace. Like the

General Hospital, this nine story

reinforced concrete frame structure col lapsed towards the east,

exhibiting localized failures at the beam-to-beam joints of each floor. The Juarez Hospital received

considerable media attention as rescue personnel continued to retrieve survivors as late as ten days after the earthquake by

tunnel Ing through the debris between the floor slabs. Heavier rigging

equipment and more rapid slab cutting

techniques -- such as an oxygen lance

-~ may have substantially reduced the time required to remove the debris

and reach possible survivors. Taller concrete buildings, such as apartments over 5 stories, were particular victims of earthquake

damage. The 14 story Nuevo Leon Complex consisted of three

Independent wings separated by

construction joints. The two

northern-most wings, each containing

196 apartments, apparently collapsed

from overturning due to column

failure on the east side of the

building.

Heavy damage was also sustained in the Juarez apartments (figure 12), just west of the Medical Center. Two eight story, and one thirteen story

reinforced concrete flat slab

structures collapsed in this area.

358
The southern half of the thirteen story

complex, separated by a non-structura

I

joint, remained standing but required

demo i i t i on

.

Perhaps the most impressive failure was

that of the two southern towers In the Consume Pino Suarez, a government

office complex Just west of the Juarez Hospital. These 14 and 21 story

buildings utilized an exterior welded box column frame and an Integral

composite slab system. These were among the few steel structures to collapse as a result of the earthquake.

Despite heavy truss floor beams very little ductility appears to have been developed due to buckling of the truss braces under reversed loading.

However, the col lapse appears to have been Initiated by column buckling in the 21 story tower, as evidenced by substantial listing to the south of the one of the remaining identical

structures.

4. POUNDING PROBLEMS

Another common cause of building damage

was drift interaction between adjacent structures (figure 13), Here two

buildings of similar height were built too closely together. Their natural period was close to the period of the earthquake allowing the structures to undergo lateral displacements

sufficient to allow them to "hammer”

each other. The structure on the left.

In this case, lost the battle for survival. This phenomena was more pronounced when buildings of unequal

height were Involved. Despite the many collapses there were also many success stories. Well designed masonry Infill frame

structures. In general, safely

dissipated earthquake energy In the shear walls, as evidenced by well developed x-cracking In the masonry.

Steel frame structures, such as the 43 story Latin American tower, recorded

little damage principally because their natural periods of vibration were significantly longer than the

charcter i st : c ground motion period for the sites on which they were

constructed. Would these towers, have performed equally well In the

unconsol Idated soil near the airport

where the natural period of the quake was much longer and the duration

sufficient to drive any structure with

a period near 4 seconds to fatal

resonance?

Many tal I concrete structures whose design met the requirements of the building code did perform well.

Given the magnitude and duration of the quake the performance of the city

as a whole — with Its one million structures -- can only be considered

Impressive.

5. SUMMARY

In summary, the primary causes of structural failures were unusually

high accelerations combined with an exceptionally long duration of shaking. The high accelerations were caused by amplification due to local

soil conditions. The coincidence of the period of ground shaking with the natural period of a large number of buildings, and foundation failures

due to a loss of soil shear capacity

led to the catastrophic damages. Field I nvest Igator

s

National Bureau of Standards

Dr. Edgar V. Leyendecker, Structural Engineer, team leader Dr. William C. Stone, Structural

Eng I neer Dr. Felix Yokel, Geotechnical

Engineer

United States Geological Survey

Dr. Thomas Hanks, Seismologist

Dr. Mehmet Celebi, Civil Engineer

6. ACK NOWLEDGEME NT The assistance of engineers and researchers from the University of Mexico (Universidad Naclonal Autonoma

de Mexico) In freely providing timely

and detailed data on this earthquake

Is gratefully acknowledged. We are also Indebted to Ambassador John

Gavin, the staff of the U.S. Embassy

In Mexico City, and the Office of Foreign Disaster Assistance of the Agency for International Development

for their support of the field

investigation.

359
Figure 1. Location map showing epicenters of main 8.1 earthc'uake and 7.5 aftershock.

COCOS PLATE NORTH AMERICAN

PLATE

MEXICO

Figure 2. Geoligic profile showing interact in between the Cocos and North American plates.

360
Figure 3. Aerial view of Mexico City looking south.

Type of

Construction Ccndition Year<1957 of CcJTStruction 1957-1976 >1976 Number of Stories

<5 6-10 11-15 >15 Total Reinforced Collapse 27 51 4 27 46 8 1 82 Concrete fram Serious 16 23 6 io 28 6 1 45 Steel frame Collapse 7 3 0 4 3 1 2 10 Serious 1 1 0 0 0 2 0 2 Flat slab/ Collapse 8 62 21 36 49 5 1 91 waffle slab Serious 4 22 18 5 26 12 1 44 Mascnry Collapse 6 5 2 11 2 0 0 13 Serious 9 13 1 22 1 0 0 23 Other Collapse 4 8 2 12 2 0 0 14 Serious 0 4 2 2 4 0 0 6 Collapse 52 129 29 89 103 14 4 210 Serious 30 63 27 40 58 20 2 120 TOlMi 82 192 56 129 161 34 6 330

Figure 4. structural damage summary.

361
362
Figure 7. Aerial view of comiBunications Center (SCT)

.

Figure 8. Acceleration records at the SCT.

363
V, l\rv]i \/

'

' ~-v'^'nAA^AAiWlM^

I I I 1 I I..I.1 1,1, L,I 1 1 1 I I I I t L l.i...L.l <, 1 t,i I ie 24 22 40 41 SO 04 T2 TIME (••cond«)

Figure 9. East - west ground motion at the SCT.

Figure 10. SCT response spectra.

364
General Hospital

Juarez Hospital

Figure 11. Examples of hospital damage.

365
Juarez

Figure 12. Examples of housing damage.

366
Pounding of equal height buildings

Pounding of unequal height buildings.

Figure 13. Building interaction.

367
JAPAN-MEXICO TECHNICAL COOPERATION

ON DAMAGE EVALUATION, REPAIR AND STRENGTHENING FOR BUILDINGS DAMAGED BY 1985 MEXICO EARTHQUAKE

BY Masaya Hirosawa(I) and Tatsuo Murota(II)

ABSTRACT local building administration offices and construction corporations. Names and

The government of Japan dispatched a team of experts to Mexico from September 30 to October 7, affiliations of members are: 1985 to negotiate a cooperation on restoration Norihisa Ishikawa Senior Coordinator,

of buildings damaged from the September 19-20, 1985 Mexico Earthquakes. A report on recommendations on damage evaluations was prepared for

the Department of Federal District of Mexico.

Mission Leader Division of Building

Standards, Bureau of Housing, Ministry of Construction

1 . INTRODUCTION . A team of experts visited the Mexico City

earthquake site during September 20 - October

Tsuneo Okada Professor, Institute of Industrial Science, University of Tokyo

7, 1985. Their mission was to negotiate

cooperation to renovate buildings damaged by

the September 19-20, 1985 earthquake. In this negotiation, it was agreed by both sides that

the Japanese government shall dispatch experts

Masaya Hirosawa Director, Planning

Division, building Research

Institute, Ministry of Construction

and donate equipments as technical cooperation

on investigation, judgment and retrofitting of damaged buildings. Based on this recognition,

a second expert team was dispatched to Mexico

during October 19 to November 22. The team was

Mizuno Yamada Manager, Design Department, Building Construction

Division, Taisei Corporation

sponsored by the Japan International Cooperation Agency. Its mission:

(1) to investigate and analyze typical

building damage in Mexico City,

Hisahiro Hiraishi Senior Research Associate, Structural Engineering

Division, Building Research

Institute, Ministry of Construction

(2) to recommend damage evaluation guidelines

and repair and strengthening methods after modifying those used in Japan,

(3) to donate equipments and documents useful

for applying these damage evaluation

guidelines and repair and strengthening

Kazuyoshi Ohshima Assistant Division Chief, Division of Building

Design, Department of Building and Repair, Secretariat of the Ministry

of Construction

methods and

(4) to recommend methods to decrease possible

damage to buildings in future earthquakes

Sh inj i Taki guch

Assistant Division Chief, Division of Supervising, ditto

in Mexico.

2 . MEMBERS

The second mission consisted of 22 experts from disciplines of earthquake engineering and building construction. They were selected from

Yuj i Naj ima Coordinator, Division of Building Design, Department

of Building and Repair, Bureau of Construction

(Kanto Area) , Ministry of Construction

the Ministry of Construction, universities,

(I) Director, Research Planning and Information Dept., Building Research

Institute, Tsukuba, Japan

Tatsuo Murota Head, Building Aerodynamic

Section, Structural Engineering Division, Building Research

Institute, Ministry of Construction

(II) Director, Structural Engineering Dept.,

BRI
Toshihide Kashima

Hideo Tsukagoshi

Toshikazu

Kawashima

Hideaki Kishida

Tadao Minami

Masaya Murakami

Junichi Sato

Toshiyasu

Yanagihara

Hiroaki Eto

Shunsuke Sugano

Kazuo Tamura

Toshiyuki Noj

i

Tsuneyoshi

Hagiwara

Research Associate, Structural Engineering

Division, ditto

ditto

ditto

Professor, Department of Science and Technology, Graduate School of Tokyo

Institute of Technology

Associate Professor, Earthquake Research

Institute, University of Tokyo

Professor, Department of Architectural Engineering,

Faculty of Engineering, Chiba University Assistant Division

Director, Building Standard

Division, Bureau of Urban

Planning, Tokyo

Metropolitan Government

Technical Director, Architecture Division, Government of Shizuoka

Prefecture Assistant Supervising

Researcher, Structural Engineering Department, Technical Research

Institute, Ohbayashi

Corporation

Chief Research Engineer, Division of Structural

Engineering, Department of Research Development, Technical Research

Laboratory, Takenaka Komuten Co., Ltd. Ohsaki Research Institute, Shimizu Construction Co.

,

Ltd. Chief Research Engineer, Technical Research

Institute, Mitsui Construction Co., Ltd. Chief Research Engineer,

Soil Mechanics and Foundation Department, Institute of Construction

Technology, Kajima

Corporation

3. TIME TABLES OF TECHNICAL ACTIVITIES

The five members in the preceding list departed

Tokyo and arrived at Mexico City on October 19. On the 21st, they held a meeting at the Department of Federal District of Mexico with

Mr. Noreha, Secretario General de Obras and Mr. Rivas, Coordinator Tecnico Operativo, where objectives of technical cooperation and

tentative schedule of technical activities were

confirmed. During the following week, the members visited the sites of damaged buildings

in central parts of Mexico City with Mr. Ortiz, Subdirector de la Coordinacidn Tecnica Operativo, who had been nominated as the DDF's senior counterpart of this mission, to select

some damaged buildings as the objects of investigation. During this period, they visited universities and consulting engineers

to collect information and held discussions to learn more about the building damages. The other 17 members arrived in Mexico on October

28. From October 30 to November 6, they investigated three damaged buildings in some

details: a 12 story office building and a 4 story school buildings. From November 7 to 12,

a 9 story apartment building was investigated.

In addition to these investigations, quick

inspection of safety and damage evaluation were

made on other 16 buildings. Microtremors were observed in 46 buildings. This work was performed under the cooperation of Technical

staffs of DDF and engineers of consultant

corporations in Mexico City. Table 1 shows the technical activities time table.

4. TECHNICAL REPORT

A technical report entitled "Recommendations on

Damage Evaluation, Repair and Strengthening for Buildings Damaged by the September 19-20, 1985 Mexico Earthquakes" was submitted to the Department of Federal District of Mexico by the Mission on Nov. 21. This report consists of

five chapters;

Chap. 1 contains objectives, members committed

and timetables of technical activities. Chap.

2 are describes the earthquakes damage evaluation guidelines for damaged buildings; guidelines for quick inspection of safety and two levels of damage evaluation, which are revi- sions of the damage evaluation guideline

developed recently in Japan based on damage

observations, strengthening practices and

various kinds of structural tests. Chap. 3 describes some methods to repair and strengthen

buildings, currently adopted for Japanese

buildings. Comments about the repair and strengthening of substructures are also contained in this chapter. Chap. 4 contains

twenty case examples of applications of guidelines and methods on existing Mexico City

buildings, four cases address damage evaluation

and repair and strengthening, and sixteen case examples address quick inspections for

369
determining the safety and level -1 damage

evaluation. Chap. 5 is a summary of recommendations to DDF from the Mission.
Table 1 Time Table of Technical Activities

Oct. Nov.

21 28 k 11 18

I Selection of Damaged Buildings

to be investigated

Collection of Informations Investigation of

#I to )j<3 Building

Investigation of

ffh Building

Micro Tremor Observation —I 1 Collection of Information

about Soil and Foundations

Analysis of Data, Report Making

o o

Meeting with DDF o o

Delevery Ceremony

371
ASPECTS OF STRONG MOTION FROM THE MICHOACAN, MEXICO,

EARTHQUAKE OF SEPTEMBER 19, 1985

John G. Anderson', James N. Brune', Jorge Prince', Enrique Mena’ Paul Bodin', Mario Onate^

Roberto Quaas^ Shri Krishna Singh^ ABSTRACT

Strong motion accelerograms from the Mexico earthquake

of September 19. 1985 provide the most complete picture to date of the damage-producing motions of a magnitude 8-1- earthquake. The success in obtaining these records was

significantly enhanced due to close cooperation between the

fields of seismology and earthquake engineering, and also between U.S. and Mexican scientists, INTRODUCTION

Strong motion accelerographs are operating in Mexico as the result of two major sources of support. The first general category is instruments installed and maintained entirely from

any of several Mexican sources for funds. The agency in Mexico which maintains all of these instruments is the Institute de

Ingenierfa (II), at the Universidad Nacional Autonoma de

Mexico. Instruments in this category include all accelerographs

in Mexico City and many digital and analog accelerographs scat- tered throughout Mexico. The second category includes the Guerrero accelerograph network, described here, and the Baja California accelerograph network (Anderson et al., 1982) has been installed with support from the U.S. National Science Foundation in a cooperative effort between II personnel and

personnel at the Institute of Geophysics and Planetary Physics,

at the University of California, San Diego. GUERRERO ARRAY

The Guerrero array consists of 30 digital accelerographs which were installed for the specific purpose of obtaining strong motion records from a large (M>8) earthquake. The array was strategically located in Guerrero and Michoacan, Mexico, m two seismic gaps, with the expectation, obtained from

seismological research, that these locations were high probability areas for such a large earthquake. Figure 1 shows a space- time diagram of Mexico to illustrate that prior to the 1985

events, the Michoacan region was a seismic gap. This was the

basis of the placement of the accelerographs in that location. Actually, prior to the September 1985 earthquakes, we were

uncertain about the seismic potential of the Michoacan region, where there were no certain prior earthquakes. In the Guerrero

gap, in contrast, there were several large earthquakes at about

the turn of the century (1899, 1908, 1909, 1911). Table 1 lists ail of the stations in the Guerrero array. The

1985 earthquakes occurred at a time when installation of the

array was about 65 percent complete. Therefore there are only

16 accelerograms for the earthquake from that array. Figure 2 shows the locations of instruments in the Guerrero array which recorded the 1985 earthquakes. Most of the

units are m the Mexican state of Guerrero, m the Guerrero

gap, but several instruments were located in Michoacan, and

were in the immediate vicinity of rupture during the September

1985 event.

'Institute of Geophysics and Planetary Physics, University of

California, San Diego, La Jolla, California, U S. A.

^Instituto de Ingenierfa, Universidad Nacional Autonoma de

Mexico, Mexico City, Mexico

MEXICO CITY ACCELEROGRAPHS

Table 2 lists Mexico City vicinity accelerographs which recorded the September earthquake and Figure 3 shows

the locations of those which were located in Mexico City in more or less free field conditions. The soil conditions in Mexico City vary widely, from relatively firm soils or volcanic flows

in the hills zone to soft clay in the lake zone as indicated in Table 2 and Figure 3. SEISMOLOGICAL CHARACTERISTICS

Before discussing the strong motion it is useful to review some seismological characteristics of the Sept. 1985 Mexico

earthquake. Several parameters are listed in Table 3. For comparison, Table 3 also shows some of the characteristics of the March 3, 1985 Chile earthquake. The Chile earthquake is important for being about the same size, and also for producing

strong motion records from near the faulting. Figure 4 is modified from Scholz (1982). It shows the

relationship of two parameters from several earthquakes which

have occurred in a strike slip environment and several which

have occurred in subduction thrust zones, such as in Mexico.

According to all purely elastic models, for a long earthquake which ruptures the entire width of the seismogenic zone the seismic moment is proportional to the product of the rupture

length, L, and the rupture width, IF, squared. The constant of

proportionality is the stress drop, which may vary from one

earthquake to another. This figure shows that the stress drop

appears to increase as the seismic moment increases. The Mexico and Chile earthquakes have been added to the figure. These two earthquakes are not anomalous based on Figure 4. The point for the Mexico earthquake is a little low (ie, stress drop appears to be a little higher than other events of the same

size), but considering the uncertainties in interpretation of rupture width which have not yet been resolved, it is consistent with the other points

In contrast to this result, Houston and Kanamori (1986) have looked at the spectrum of teleseismic E-waves in the fre- quency range of 0.05 to 1 Hz. While these spectra are similar

at frequencies less than 0.1 Hz, from 0.1 to 1.0 Hz the Mexico

earthquake spectra are depleted relative to Chile and relative to

a worldwide average. The lower amplitudes in Mexico appear

to correlate with a less complex rupture process than that found

for the Chile event. STRONG GROUND MOTION

Levels of Peak Motions

Table 4 lists all peak accelerations, peak velocities, and

peak displacements for the main shock from the Guerrero array Table 5 lists peak accelerations, velocities, and displacements which are available from the Mexico City vicinity. Figure 5 shows peak accelerations which were recorded in the Chile and

the Mexico earthquakes. The peak values in the Mexico earth- quake are much less scattered than m Chile, and except for the

points in Mexico City, the Mexico points almost seem to pro- vide a lower bound to the Chile points. These peak

accelerations are at a higher frequency than the frequencies which were studied by Houston and Kanamori (1986), but if it is allowable to extrapolate between the two frequency bands.

372
the lower peak accelerations may be consistent with the telese- ismic results. There is however one other factor which is important in evaluating these peak accelerations, namely site effects. The

site conditions in Michoacan and Guerrero are uniform. All stations were placed on the most competent rock outcrop consistent with the target location. Only one station (El Cayaco) was not sited on rock. Compressional wave velocities measured

from rock samples taken at many of the sites have a median

value of about 4 km/sec, with the variation among sites sampled to date ranging from a low of 1.6 km/sec to a high of 5.6 km/sec. The high value in this preliminary set was obtained from the site at La Villita. Samples from the other stations above the source zone are not completed yet, but the rock samples are similar to others which have measured about 4 km/sec.

In contrast to this situation, the site conditions in the Chilean data set are not uniform, and are not yet as well docu- mented. Conditions described in Bolt and Abrahamson (1986)

are evidently obtained from geological maps of the area rather than field inspection. Some of the accelerograms which are nominally listed as coming from rock sites show apparent ringing, probably caused by local site conditions. Therefore it is likely that to some extent the lower levels obtained near the

faulting in Mexico are the result of differing site conditions. From the Mexican data alone, there is clear evidence that the site conditions are important. The accelerations in Mexico

City were unambiguously amplified by local site conditions. In addition, an II accelerograph, Zacatula 4.3 km from the La Villita accelerograph recorded peaks in acceleration of 1.5 to 2.5 times those recorded at La Villita. Peak velocities at Zacatula were also amplified by a similar amount. The Zacatula station

is situated on a river terrace deposit of probable Quaternary

age. The deposit is an indurated cobble conglomerate with a high density of cobbles ranging in size mostly from 5 to 25 cm. Appearance of Accelerograms

Figure 6 shows the north-south component of all accelerograms from the Guerrero array. Figure 7 shows the east-west acceleration from those Mexico City sites which are shown in Figure 3. At the stations above the faulting (Caleta de Campos, La Villita, La Union) the duration of strong shaking is fairly long (over 30 seconds) but at stations farther to the

southeast, the duration is shorter. This is because the rupture on the fault propagated to the southeast during the event, so

that the arrivals in that direction are compressed in time. In Mexico City, the durations are much longer than in the source

region. At the station CDA and SCT, the strong shaking lasted over 60 seconds (only the most significant part of these records

is shown), and the total duration of these accelerograms exceeds 180 second. Anderson et al. (1986) have demonstrated that above the

fault, the accelerograms can be integrated to obtain a per- manent offset. The offset obtained is consistent in magnitude

and direction with that expected from the earthquake mechanism. Vertical offsets are also apparently consistent with observed uplift along the coast. Figures 8 to 13 show a selection of accelerograms and

response spectra from the Guerrero array. These records have

been filtered at low frequencies, with the cutoff in the vicinity of 0.15 Hz, so the static displacements are not visible. Figures

14 to 16 show the same information for a selection of records from the Mexico City vicinity. Because ot the unprecedented building damage in Mexico

City for the earthquake and because the damage was in large

part caused by amplified 2 second energy, an important question is whether or not the earthquake source energy at 2 second

period was anomalous compared, for example, with previous

large Mexico earthquakes. A survey of the velocity and dis- placement records outside of the Mexico City basin (Figures

8 — 13) shows obvious high amplitudes at these periods, e.g. especially records from Caleta de Campos, La Villita, Papanoa,

Teacalco, and CUIP. Some records from the coast show local maxima in the response spectra at 2 seconds (e.g. Caleta de Campos and La Villita) but at all stations closer than El Cayaco

(Figure 2) the response spectra also attain even higher values

of periods shorter than 2 seconds. The highest peaks with velocity response spectra near 2 second period on the near-source

stations Caleta de Campos and La Villita (~60 cm/sec) for zero damping are smaller than the response spectra peaks near

the same period for 1971 San Fernando (Pacoima Dam > 150 cm/sec, near 1.5 second period) and 1940 Imperial Valley (El Centro >110 cm/sec, near 2.0 second period) (EERL 72-80,

73-81), Thus there does not appear to be any clear indication of anomalously high source energy near 2-seond period, but a

final conclusion awaits a more thorough comparison with other Mexico earthquakes. Earth structure effects (e.g. energy focusing) also might have been partly responsible for the unusual damage for the earthquake. CONCLUSIONS

Earthquake hazards are distributed widely around the

world, and the need to measure the strong motion in the vicinity of strong earthquakes is an international problem. The success in obtaining accelerograms from the epicentral area of the Mexico earthquake is a result of application of seismological information about the relative likelihood of earthquakes, and

international cooperation to put the instruments in one of these

locations. ACKNOWLEDGMENTS

This research was supported by National Science Foundation grants CEE82- 19432 and ECE- 13489. REFERENCES

(1) Anderson, J.G., J.N. Brune, J. Prince, and F.L. Vernon III (1983). Preliminary report on the use of digital strong motion recorders in the Mexicali Valley, Baja California,

Bull. Seism. Soc. Am., 73. 1451-1468.

(2) Anderson, J.G., P. Bodin, J. Brune, J. Prince, and S. Singh

(1986). Strong ground motion of the Michoacan. Mexico

earthquake, (invited paper Science, accepted for publciation).

(3) Bolt, B, And N. A. Abrahamson, Seismological features, in Earthquake Engineering Research Institute report on the March 3, 1985 Chile earthquake (in press), (1986)

(4) EERL, Analyses of strong motion earthquake accelero- grams, Vol. Ill - Reponse Spectra, Part A - Acclerograms

IIA0011IA020, Earthquake Engineering Research Laboratory, Calif. Institute of Technology, Pasadena, Calif. 272

pp., (1972).

(5) EERL, Analyses of strong motion earthquake accelero- grams, Vol. Ill - Reponse Spectra, Part C - Acclerograms

1IC041IIC055, Earthquake Engineering Research

Laboratory, Calif. Institute of Technology, Pasadena. Calif.

• 191 pp., (1973).

(6) Houston, H. and H. Kanamori, Source characteristics of the 1985 Michoacan, Mexico earthquake at short periods, (Geophys. Res. Letters, in press), (1986).

373
(7) Institute de Ingenien'a de la Universidad Nacional Autonoma de Mexico, El temblor del 19 de septiembre de

1985 y sus efectos en las construcciones de la Ciudad de Mexico, Informe Preliminar Institute de Ingenien'a, UNAM, Mexico City, (1985),

(8) Scholz, D. H., Scaling laws for large earthquakes, consequences for physical models. Bull. Seism. Soc. .4mer. 72. 1- (9) 14 (1982). Sykes, L. and R. Quittmeyer (1981). Repeat times of great earthquakes along simple plate boundaries, in D.W. Simpson and P. G. Richards, eds.. Earthquake Prediction: an

International Review. 217-247.

Table 1. Accelerograph Stations in the Guerrero Array Which Were

Installed Prior to the September 19, 1985 Earthquake

Location Instrument

Station Lat.°N Long.°W Type S/N Geology

Caleta de Campos 18.0727 102.7552 DSA-1 261 metavolcanic La Villita 18.0475 102.1840 DSA-1 260 gabbro La Union 17.9824 101.8054 DSA-1 259 Metavolcanic

Zihuatenejo 17.6030 101.4550 DCA333 106 tonalite Papanoa 17,3278 101.0399 DCA333 111 weathered plutonic

El Suchil 17.2258 100.6418 DCA333 109 Plutonic Atoyac 17.2113 100,4309 DSA-1 256 weathered plutonic

El Paraiso 17.3444 100.2145 DSA-1 257 plutonic

El Cayaco 17.0452 100.2664 DSA-1 258 Coyuca 16.9967 100.0900 DCA333 107 schist/gneiss

El Ocotillo 17.0378 99.8749 DCA333 110 weathered gabbro

Xaltianguis 17.0950 99.7201 DSA-1 253 granodiorite

El Ocotito 17.2500 99.5106 DCA333 104 weathered plutonic La Venta 16.9129 99.8159 DSA-1 252 granitic gneiss Cerro de piedra 16.7692 99.6326 DCA333 112 weathered gneiss Las Mesas 17.0070 99.4565 DSA-1 254 weathered granitic/gneiss San Marcos 16.7760 99.4077 DSA-1 255 granitic gneiss Las Vi gas 16.7560 99.2359 DCA333 108 plutonic Tonalapa 18.0975 99.5594 DSA-1 249 met. shale Teacalco 18.6174 99.4528 DSA-1 248 volcanic tuff
Table 2. Accelerograph Slatons in the Vicmiiy of Mexico City

Station Code Lat,“ N

Long= W Type/ NO

Accelerograph Comp, Nat. Freq. Damp G

Sue Description

Cd. Univ, D.F CUOl 19,330 DCA333 SOOE 33.0 0.64 Rock basalt. 1st lloor 99.183 -121 VERT 31.0 0,66 of 3-story Idel mam

N90W 32.0 0 70 Ibldg. Hilly zone

Idei Patio CLIP 19.330 DCA310 NOOE 31,0 0.73 Rock, basalt. 99.183 -154 VERT 31.3 0.66 Free field. N90W 31.4 0.68 Idel yard. Mesa Vibra- CUMV 19,330 DCA333 SOOE 31.0 0.70 Rock, basalt. dora C.U. 99,183 -145 VERT 31.0 0.66 Shaking table. N90W 30.0 0.75 Free field. Sismex SXPU 19.043 DCA310 NOOE 32.0 0.71 Hard soil, sand and

Puebla 98.212 -148 VERT 32.0 0.77 gravel. Free field. N90E 31.0 0.74 Puebla Valley. Tacubaya. D.F TACY 19,403 DCA333 SOOE 30.0 0.70 Hard soil. 99.194 -168 VERT 30.0 0,70 Free field. N90W 30.0 0.70 Hill zone, D F

Sismex SXVT 19,358 DCA310 NOOE 30,4 0,66 Soft soil, clay. Viveros 99,171 -133 VERT 32.0 0.80 Free field. N90W 330 0.78 Transition zone. D F, C de .Abasios CDAF 19 368 DCA333 SOOE 31,0 0.70 Very soft soil. Frigonfico 99 088 -114 VERT 31.0 0.70 Free field. Texcoco N90W 31.0 0.70 lake bed zone. D F. C. de .Abastos CDAO 19,368 DCA333 SOOE 31.0 0.70 Very soft soil, clay. Oficina 99,088 -115 VERT 31.0 0.70 1-story bldg. Texcoco N90W 31.0. 0.70 lake bed zone. D F. Secretaria de SCTI 19.393 DCA333 SOOE 300 0.75 Very soft soil, clay. Comumcaciones 99.147 -144 VERT 30.0 0.66 Free field. Texcoco

y Transportes N90W 30.0 0.70 lake bed zone, D F. TIahuac TLHB 19.279 SMAl NOOE 18.3 0.60 Soft soil. clay. Bombas 99 008 -4590 VERT 19,0 0.60 Free field. Xochimilco N90W 18.5 0,60 lake bed zone

TIahuac TLHD 19.293 SMAl NOOE 18.3 0.60 Soft soil. clay. Deportivo 99 035 -4591 VERT 19.0 0.60 Free field. Xochimilco N90W 18.5 0.60 lake bed zone

Table 3. Characteristics of Mexico and Chile Earthquakes

Event Michoacan

Date Sept. 19, 1985 Time 13;17

-V/s 8.1 Mq (dyn-cm) 0.9-l.Sx 10^8

L (km) 170 W (km) 50

Michoacan Chile

Sept 21, 1985 Mar. 3, 1985

01:37 22:46

7.5 7,8

l,4x 10^^ l.lx 10^8

70 125

40 90

375
Table 4. Strong Motion Data from the Guerrero Array

Sept. 19, 1985 Mexico Earthquake (,4/5=8.!)

Station R 1 R2

(1) Trigger

(2) Dur.

(3) Comp.

(4)

,4 max

(5)

l^max (6) Dma,

(7)

l^max (8) Dma,

(9) Caleta 22.7 0.0 13:17: 49.4 N 138 25.9 16.70 5.73 de 0.0 W 141 19.5 12.64 3.22 Campos 0.0 DN 89 16.9 7.11 1.88 0.0 13:18: 25.6 N 40 4.1 4.4 4.04

0.0 W 51 4.8 3.2 5.96 1.18 0.0 DN 24 3.3 2.0 2.91 0.88 La Villita 43.7 0.0 13:18: 64.0 N 125 25.9 16.11 6.67

0.0 W 122 10.51 3.77

0.0 DN 58 11.0 4.72 1.21 La Union 84.0 0.0 13:18: 62.8 N 166 28.0 20.34 6.95

0.0 W 148 14.4 11.70 4.16

0.0 DN 129 17.7 11.72 3.89 Zihuatenejo 134.6 8.0 13:18: 71.8 N 103 20.8 34.9 15.86 5.96

8.0 W 161 17.9 17.2 18.34 4,62

8.0 DN 105 14.6 22.1 10.23 3.80 Papanoa 187.9 63.0 13:18: 59.5 N 162 14.0 21.2 967 2.85 63.0 W 117 7.7 12.9 6.13 1.57 63.0 DN 84 10.4 14 6 8.38 1 81

El Suchil 230.4 112.0 13:18: 60.0 N 102 16.0 17.5 11 61 2.95 112.0 W 81 7.1 7.9 6.36 1.49 112.0 DN 51 6.8 9.2 5,96 1.16 Atoyac 251.1 135.0 13:18: 42.2 N 53 8.0 139 5.44 2.04 135.0 W 59 4.6 5.5 3.67 0,89

135.0 DN 60 8.3 8.6 6 68 1.61

El Paraiso 266.3 13:18: NW

DN

El Cayaco 275.0 152.0 13:18: 25.6 N 41 5.2 10.0 3.75 1.31 152.0 W 48 3.5 2.3 2.91 0.68 152.0 DN 24 5.4 9.0 3.81 1.32 Coyuca 294.1 175.0 13:18: 38.8 N

42’ 9.0 12.4 7.82 2.83 175.0 W 35 3.6 5.3 3.16 1.19 175.0 DN 19 4.2 7,8 3.33 0.91 La Venta 324.3 206.0 13:18: 10.2 N 18 5.5 2.7 4.30 1.29 206.0 W 21 3.6 1.5 2.62 0.88 206.0 DN 16 3.3 1.5 3.20 0.98 Cerro de 348.9 225.0 13:18: 25.5 N 27 5 9 9.2 2.59 0.69

Piedra 225.0 W 15 2.52 0.57 225.0 DN 13 3.8 7,1 3.07 0.79 Las Mesas 354.9 237 0 13:18: 14.1 N 22 4.6 5.1 1.90 0.51 237.0 W 18 3.2 3.3 2.84 0.66 237.0 DN 19 4.5 3.3 3.85 0.98

Xaltianguis 325.3 207.0 13:18: N 25 207.0 W 18 207.0 DN 20

El Ocotito 340.2 223.0 13:18: N 49 7 65 2.92 223.0 W 54

223.0 DN 21 3.72 1.48 Teacalco 332.5 244.0 13:18: 38,4 N 49 9.4 12.6 7.38 3.68 244.0 W 24 5.4 6.0 4,74 1 .36 244 0 DN 27 5.8 10.3 4.41 1.62

El Ocotillo 312.7 San Marcos 370.0 Las Vi gas 387.5 Tonalapa 318.3 NOTES

1. Epicentral distance (km), based on epicenter at 18.182° N, 102.573° W, origin time 13:17:47.8.

2. Closest distance to outline of aftershock zone (km)

3. Origin time of accelerogram.

4 Duration (seconds) of accelerogram.

5. Orientation, while nominal, is expected to be accurate.

6. Peak acceleration (cm/sec^).

7. Peak velocity (cm/sec) This is obtained from an integration which did not employ a high-pass filter at low frequencies, made possible because the accelerograph is digital. There are some uncertainties in this, however, and these

values are preliminary

376
8. Peak displacement (cm). Obtained from the unfiltered velocities. The same notes apply. For some records, the peak velocity and peak displacement are reduced merely because the instrument did not run long enough to recognize their presence.

9. Peak velocity (cm/sec). To obtain these values, a high pass filter with a corner at about 0 015 Hz has been applied. The much lower values present in this case result because there are very strong low frequency signals period about

20 seconds) which have been filtered out from the records. The low frequencies are physically meaningful, and can

be shown to correlate from one accelerogram to the next along the coast.

10. Peak displacement (cm). Filtered records. See note 9.

377
TABLE 5. Strong Motion Data from Mexico City and Vicinity

Sept. 19, 1985 Mexico Earthquake (A/s =8.1)

Station R 1 (1) R2

(2) Trigger(3) Dur.

(4) Comp.

(5) (6) Pmax

(7) Omax

(8) Pmax

(9) Omax

(10) Teacalco 332 244 13:18: 38.4 N 51.3 9.4 12.6 7.38 3.68 244 W 24.7 5.4 6.0 4.74 1.36 244 DN 27.1 5.8 10.3 4 41 1.62 CUOl 379 289 60.0 S 28.1 10.20 5.54 W 33.5 9.38 7.19 UP 21.6 8.23 6.65 CUIP 379 289 60.0 N 31.7 10.25 6.17 w 34.7 9.37 7.71 UP 21.9 7.99 6.61 CUMV 379 289 60.0 s 37.4 9.19 5.70 w 38.8 11.01 4,45 UP 20.1 8.40 4.57 SXPU 469 379 85.3 N 29.6 7 18 3.10 E 32.6 6.63 2.65 UP 15.9 4.06 1.67 TACY 380 290 156.0 S 34.4 14.29 11.96 w 33.2 9.81 8.62 UP 19.2 9.29 7.61 SXVI 381 291 60.0 N 44.1 11.48 9.10 E 42.4 12.20 7.47 UP 18.2 5.84 6.96 CDAF 389 299 60.0 S 80.5 24.85 15.02 w 94.6 37.57 18.88 UP ' 27.2 8.90 8.17 CDAO 389 299 180.0 N 69.2 34.98 25.02 E 80.4 41.86 24.67 UP 35.7 11.34 8.92 SCTl 385 295 180.0 s 98.0 38.74 19.12 w 167.9 60.50 21.94 UP 36.6 9.00 7,58 TLHB 394 304 150.0 N 135.9 64.10 36.61 w 106.7 44.61 39.29 UP 24.0 9.39 6,78 TLHD 392 302 150.0 N 117.7 34 90 20.76 w 111.6 36.06 22.07 UP 59.3 17.01 6.87 NOTES

1 Epicentral distance(km). based on epicenter at 18.182 N, 102.573 W, origin time 13:17:47.8.

2. Closest distance to outline of aftershock zone (km). Taken to be 90 km less than distance to epicenter.

3. Origin time of accelerogram.

4 Duration (seconds) of accelerogram.

5. Orientation, while nominal, is expected to be accurate.

6. Peak acceleration (cm/sed).

7. Peak velocity (cm/sec). This is obtained from an integration which did not employ a high-pass filter at low frequencies, made possible because the accelerograph is digital. There are some uncertainties in this, however, and these values are preliminary

8. Peak displacement (cm). Obtained from the unfiltered velocities. The same notes apply. For some records, the peak veloccity and peak displacement are reduced merely because the instrument did not run long enough to recognize their presence.

9. Peak velocity (cm/sec). To obtain these values, a high pass filter with a corner at about 0.015 Hz has been applied. The much lower values present in this case result because there are very strong low frequency signals (period about

20 seconds) which have been filtered out from the records. The low frequencies are physically meaningful, and can be shown to correlate from one accelerogram to the next along the coast.

10. Peak displacement (cm). Filtered records. See note 9

378
YEAR

Figure 1. Top: map of Mexico with epicenters of large historical earthquakes identified by year, and depth in parentheses when event is known to be deep. Bottom: Space-time diagram in which epicenters are projected down and plotted at year of occurrence. Deep events and events located far inland and thus not clearly related to subduction are deleted. Larger events are highlighted. Guerrero gap and former Michoacan gap are illustrated.

379
1985 Guerrero accelerograph array

Figure 2. Map of coastal Mexico, epicenters and aftershock zones of 1985 events, and locations of strong motion sta- tions in Guerrero array on September 19, 1985. Short dashed lines show 'imits of aftershocks of large earthquakes in this region since 1951. Peak accelerations, in cm/sec^ are given for each station, for the north, east, and vertical components, respectively, in parentheses.

380
LAKE ZONE

ACCELEROGRAPH

SEVERELY DAMAGED BUILDING COLLAPSED BUILDING ZONE WITH MANY COLLAPSED

1 AND 2 STORY HOUSES

(BRICK AND ADOBE )

Figure 3. Mexico City map showing free-field accelerograph stations, generalized soil classification, and sites ot worst

building damage from Instituto de Ingeneria (I985) preliminary report. Peak accelerations (cm/sec^) for the north,

east, and vertical components, respectively, are given in parentheses by each station.

381
Figure 4. Plot of log LW^ versus log Mq for the large interplate earthquakes from the data set of Sykes and

Quittmeyer (1981). The lines of slope 1 are constant stress drop lines, assuming C=0.6 for the thrust events, and 0.3 for the strike-slip events (after Scholz, 1982).

382
800

700 -

cvj 600 0

d>

CO \

1

500

p

400

<

(T

LxJ 300

200

100

MEXICO SEPT. 19, 1985 Mg

CHILE MAR. 3, 1985 Ms =

•

•

•id

r * t §•,

50 100 150 200 250

DISTANCE TO NEAREST AFTERSHOCKS (km)

Figure 5. Peaks of horizontal components of acceleration plotted against distance outside the boundary of the aft- ershock zone (Figure 2).

8.1 •

7.8 °

300

383
N-S ACCOLRATION

30^

0

1||) ^ 20 _ 30 40 50 60 70 30

30 60

la union

70 30

50 60 ZiHUATENEJO

TO 90

00|- 0 ^0 1 . M 1. i 50 60

PAf'ANOA

70 80

60|°

Ij.

50 60

EL SUCMIL

70 30

50 60 ATO/AC

70 80

.0

04 •

50 60

EL PARAISO

70 so

40 1 0 iL 50 60

FL CATACO

70 80

|(“—

50 60 COYUCA •“ 70 80

int l\ t 20 30 40 4

50 60

lA Vt.NTA 70 30

20 r 0 ijp 20 30 40 50 60 CEPRO DE PltDRA

70 80

20h 0 j 10 20 30 40 “F

50 60

las IdLSAS 70 30

0 10 20 30 40 50 60

<ALT1ANGUIS

70 30

oefO ’0 2C 30 40

. .

SO 60

CL OCOT.TO 70 30

50 60

TEACA4.C0 70 80

0 10 20 30 40 50 60 70 80

Figure 6. North-south component of acceleration for stations above the aftershock zone. Vertical separation of traces

IS proportional to separation of the projection of stations onto the trench. Time, Tq, is the origin time of the earth- quake. The clock correction at Caleta de Campos is uncertain.

384
EAST-WEST ACCELERATION

0 - UNAM

VIV

Figure 7. Most significant one-minute segments of the east-west acceleration recorded on the free-field accelerographs

in Mexico City. Complete accelerograms are longer 100 sec of motion preceded the segment of the SCT record shown

here. No time correlation exists among these traces.

385
Figure 8. East west acceleration, velocity, displacement, and pseudo relative velocity response at Caleta de Camoos

September 19, 1985

UNCORRECTED ACCELERATION

0 10 20 30 40 50 60 CORRECTED AND FILTERED RECORDS

LA VILLITA N90E

Figure 9. East west acceleration, velocity, displacement, and pseudo relative velocity response at La Villita, Sep- tember 19, 1985

386
UNCORRECTED ACCELERATION

A

(gal) 1

_

o

o

o

o

o

^

*100

O’ 0

0 10 20 30 40 50 60 CORRECTED AND FILTERED RECORDS

^ -100

I? 10 - ^ E 0 ^ -5 > _io

0 10 20 30 40 50 60

4 ? 2

0 "^-2

0 10 20 30 40 50 60

0 10 20 30 40 50 60

LA UNION N90E

Figure 10. East west acceleration, velocity, displacement, and pseudo relative velocity response at La Union, Sep- tember 19, 1985

Figure 11. East west acceleration, velocity, displacement, and pseudo relative velocity response at Papanoa, Sep- tember 19, 1985

387
UNCORRECTED ACCELERATION

0 10 - 20 30 40 50 60

0 10 20 30 40 50 60

COYUCA N90W

Figure 12. East west acceleration, velocity, displacement, and pseudo relative velocity response at Coyuca, September

19, 1985

UNCORRECTED ACCELERATION

0 10 20 30 40 50 60 CORRECTED AND FILTERED RECORDS

TEACALCO -N90E

Figure 13. East west acceleration, velocity, displacement, and pseudo relative velocity response at Teacalco, Sep- tember 19, 1985

388
UNCORRECTED ACCELERATION CUIP NOOE

Figure 14. East west acceleration, velocity, displacement, and pseudo relative velocity response at Instituto de

Ingeni eria patio, September 19, 1985

UNCORRECTED ACCELERATION

0 40 80 120 160 CORRECTED AND FILTERED RECORDS

0 40 80 120 160

CDAO NOOE

Figure 15. East west acceleration, velocity, displacement, and pseudo relative velocity response at C. de Abastos

Oficina, September 19, 1985 389
Figure 16. East west acceleration, velocity, displacement, and pseudo relative velocity response at Secretaria de Comunicaciones y Transportes September 19, 1985

390
SEISMIC DAMAGE TO POI^T FACILITIES BY THE 1985 CHILE EARTHQUAKE

AND

IMPROVEMENT OF SEISMIC SAFETY OF THE DAMAGED PORTS

BY

Haj ime Tsuchida and Setsuo Noda

ABSTRACT

The March 3, 1985 earthquake (M-7.7) in the middle of Chile caused serious damage to facilities in the Port of Valparaiso and the Port of San Antonio. This paper presents the seismic damage to these facilities and the investigation of the causes of damage based on

damage reconnaissance and stability analyses.

Damage reconnaissance and planning for restoration work were performed as a technical coope- ration project between Chile and Japan. This cooperative project is described.

1. INTRODUCTION

On March 3, 1985, an earthquake of magnitude

7.7 occurred in central Chile; it resulted in lines lost and damaged engineered structures.

The Port of Valparaiso and the Port of San Antonio were affected by severe ground shaking; many port facilities suffered serious damage. Cargo handling capacities of both ports were

greatly affected. Major damage took place on quaywalls and cranes. However, some quaywalls

suffered no damaged even though they were

located in the vicinity of other damaged ones. This report presents a description of the damage to port facilities and the stability

analyses used to assess damaged quaywalls.

The Japanese government sent a team of Japanese

experts to chile. Cooperation with Chilean

port engineers for evaluating the stability of

the damage facilities was expanded as the technical cooperation effort focus on the restoration plan of the damaged facilities.

3. EARTHQUAKE AND GROUND MOTIONS

The location of the focal area is shown in Fig.

1 (see locations of the Port of Valparaiso and

the Port of San Antonio) . Event data of the earthquake are listed in Table 1. The seismic

intensities in the Modified Mercalli scale in Valparaiso were 7 to 8 and those in San Antonio

were 8. The maximum accelerations around

Valparaiso and San Antonio are listed in Table

2 (1). The Accelerographs used at the sites were the SMA-1 accelerographs with a natural period of 0.03 second.

Port and Harbour Research Institute, Structures

Division, Ministry of Transport.

3 . 1 Damase in Port of Valparaiso

3.1.1 Outline of the Port

The Port of Valparaiso is the largest seaport

in Chile and is the center of foreign trade for exporting fruits and copper and for importing

industrial products. The socioeconomics

activities of Chile are centered in the Santiago region approximately 100 km east of

the port. The port consists of mooring

facilities for 10 berths, a breakwater, 14 warehouses, and 31 wharf cranes. A plan view

of the port is shown in Fig. 2; the berth

numbers are indicated.

The mooring facilities were completed between

1915 through 1932. The seismic resistance of

these facilities structure were not taken into consideration. The structures for eight berths

are gravity type quaywalls using concrete

blocks (Berth Nos. 1 through 6 and 8) and concrete caissons (Berth No. 7). Baron Pier

(Berths Nos. 9 and 10) consists of reinforced

concrete slab-beams and concrete foundation

piers. Diameters of the foundation piers are

4.0 m. Similar concrete foundation piers also were used in the front part of the gravity type quaywall of Berth No. 4 to sustain the seaside

crane railing. The breakwater was constructed

with the concrete blocks founded on the rock

mound. The breakwater extends offshore to a water depth of 45 m. To better understand the seismic stability of

the facilities at the Port of Vaparaiso; the quaywalls at the Port of Valparaiso were damaged by the 1971 earthquake and the port operations were affected for many months (2). In particular, the quaywalls for Berths Nos. 4 and 5 are tilted with a maximum horizontal displacement of 5 cm at the ground level and a subsidence of the backfill up to 40 cm. Damage

to the quaywalls for Berths Nos. 6 and 7 were

observed. The displacements were up to 30 cm.

3.1.2

Subsoil Conditions According to the topographic survey and to the borings, the original sea bottom existed at about -10 m near the land side end of Berths

Nos. 6 and 8 and at about -35 m at Berth No. 7. Therefore, the original sea bottom is sloping

steeply towards offshore.

391
The surface layer of the area, covering

approximately Berths Nos. 4 through 8, is medium fine sand or silt having N-value of SPT about 20 and its thickness is 5 to 10 m. The surface layer at the area around Baron Pier is soft silty soil about 5 m thickness. A sand or gravel deposit lies under the surface layer and

this underlying layer has N-value of SPT not

less than 40 and it has been considered as the bearing stratum for the port facilities. The base rock formation was found at depth between

about -10 m to -17 m. As an example, the boring log at the Berth No. 1 is shown in Fig.

3.3.1.3 Damage to Mooring Facilities

The damage to the mooring facilities and to the adjacent yards were severe than the damage to the buildings. The major types of damage were:

i) sliding of the concrete blocks and the caissons (gravity elements) toward the sea, ii) tilting and settlement of quaywalls, and iii) settlement, cracking, and opening of the joints

of the apron pavements . The damage is shown in Figs . 4 through 6 . The damage to the mooring

facilities is summarized in Table 2. The amount of horizontal sliding of quaywalls

toward the sea are different depending on the berths: less damage at Berths Nos. 1 through 3 and heavier damage at Berths Nos . 4 through 8

However, the sliding is in decimeter scale. Tilting was observed at Berths Nos. 5 and 7. As shown in Photo 1, the crane railings and the railway tracks deformed and. settled. The settlement was observed in the berth aprons and

the back-of-port yard in decimeter scale. The copeline displacements were clear at berths

Nos. 4, 5, 6, and 8. The displacement at Berths No. 5 is shown in Photo 2. The copeline

displacement towards the sea was measured as 170 cm maximum, which included the displacement

caused by the 1971 earthquake.

3.1.4 Damages to Wharf Cranes

Damage to wharf cranes at Berths Nos. 6 through

8 was severer than at Berths Nos. 1 through 3. About half of the wharf cranes were damaged at

the portal legs and the rings. Damage to the

legs was caused by local buckling from large compressive stress. Some damage to the legs might have been caused by cargo striking the

legs. Derailing of land side wheels and buckling of the land side truck frames were observed at most of the wharf cranes at Berths

Nos . 6 through 8 . No damage occurred to the heavy cargo handling crane at Berth No. 4 and

to the cranes at the Berths Nos. 9 and 10. A

wharf crane at Berth No. 7 collapsed.

3 .

Damage at San Antonio Port

3.2.1 Outline of the Port

The Port of San Antonio shares, with the port

of Valparaiso, the role of chief seaports of

the country. Major cargos through the Port of San Antonio are cooper for exportation and wheat for importation. The cargos through the port sustain the socioeconomic activities in

the Santiago region. A plan view of the port

is shown in Fig. 7; the berth numbers are indicated. At the time of the earthquake, the port consisted of mooring facilities for 7 berths, a tanker berth, 2 breakwaters, 3 warehouses , and 15 wharf cranes

The structures of the mooring facilities are gravity type quaywalls at Berths Nos . 1 and 2

steel sheet pile bulkheads at Berths Nos. 3 and

4, and open type wharves at Berths Nos. 5 through 7. The mooring facilities at the port

also suffered damage from the 1971 earthquake.

The quaywall structures at Berths Nos. 1 and 2 suffered heavy tilting experiencing a maximum

displacement of over 60 cm at the ground level and subsidence up to 60 cm at the apron and the yard behind. Longitudinal cracks in the yard behind Berths Nos . 1 through 3 also were

observed. The warehouses were heavily damaged

and had to be demolished. A group of undamaged

steel silos were disassembled for safety

purposes

3.2.2 Subsoil Condition

The geological formation of the port area consists of 2 layers. The upper layer is divided into two sub-layers. The upper sub- layer extends uniformly over the port area and consists of fine sand deposit. The lower sub- layer consists of clayey deposits. The clayey

soil of the lower sub- layer is dominant at the onshore side of the port area, while the sandy

soil of the upper sub- layer is dominant at the offshore side of the port area near the breakwater. The upper sub -layer of the sands

has N-value in SPT 20 and is considered as sediments of littoral and drift from the Maipo

River. The lower layer is composed of dense

sand or gravel. From an engineering point of

view, this is the bearing layer for the port

facilities structural foundation. Its hardness

is not less than 50 N-value. The layer was

found at the depth of approximately -19 m. In

the port area, the depth of the boundary

between the two layers varies. This implies marine abrasion in the past. As an example,

the boring log at Berth No. 1 is shown in Fig.

8.3.2.3 Damage to Mooring Facilities

The damage to mooring facilities at this port

was heavier than at the Port of Valparaiso.

392
Complete collapses of the quaywalls with severe subsidence of the apron took place at the Berths Nos. 1 and 2. Some cranes were partly

damaged. Also other mooring facilities experi- enced some damage . The damage is summarized in Table 3. Figs. 9 through 11 and Photos 3 through 7 show the damage at the Port of San Antonio

.

3.2.4 Damage to Wharf Cranes

The damage to wharf cranes at Berths Nos . 1 and

2 was heavy. As shown in Photo 6, all 7 of the wharf cranes experienced severe damage or collapse. At Berth No. 5, however, there is no damage to the wharf cranes . At Berths Nos . 6 and 7 , no damage to wharf cranes were reported

although it appears that the wharf cranes may

experience difficulty operating due to rail deformation.

4. STABILITY ANALYSES AND DAMAGE CAUSES

4.1 Stability Analyses

For planing the restoration work, it is important to analyze stability and stresses of

the structures during the seismic ground

motions and to learn how the damage was caused.

The stabilities of the mooring facilities and

the cranes were analyzed by the seismic coefficient method which is currently being used for earthquake proof design of many structures.

Details of the method is available in many

references (3). In the analyses, the factor of safety for the stability of the structures were calculated for the seismic coefficients of sev- eral different values, to find the condition

which gives the factor of safety of unity. This condition may be considered as the critical condition. No loading on the apron of the wharf and the crane were considered in these analyses

.

4.2 Mooring Facilities at Port of Valparaiso

The quaywalls at Berths Nos. 1 through 8 became

the critical condition for the seismic coefficients (from 0.09 to 0.15), depending on the berth. According to the empirical formula

about the relationship between the seismic coefficient and the peak ground acceleration,

the seismic coefficient 0.09 and 0.15 corre- spond to peak ground acceleration of 0.09 g and

0.15 g, respectively (4). One of the peak

ground accelerations recorded near the Port of Valparaiso was 0.29 g (Table 2). It should be noted that the formula gives somewhat conservative seismic coefficients. Even when using the conservative coefficients, the magnitude of the earthquake ground motions exceeded what the quaywalls could withstand.

4 . 3 Mooring Facilities at Port of San Antonio

The quaywalls at Berths Nos. 1 and 2 became the critical condition for sliding when the seismic coefficient was 0.07, and for overturning at 0.11. Similar to the quaywalls in the Port of Valparaiso, it may be concluded that the cause

of damage to the quaywalls at Berths Nos . 1 and

2 was from intense seismic ground motions. On the sheetpile bulkhead at Berth No. 4, the seismic coefficient 0.17 corresponds to the critical condition. This bulkhead suffered no evident of structural damage.

The critical condition on the L- shaped

retaining wall behind the open- type wharf at Berth No. 5 occurred when the seismic coefficient is 0.17.

In the Port of San Antonio, evidences of liquefaction were found at several locations.

However, the degree of liquefaction influence

to the damage of structures in this port is not year clear. Soil testing for assessing lique- faction potential is under way. The influence

will be studied in detail after completion of testing.

4 . 4 Wharf Cranes

The wharf cranes at the Port of Valparaiso and

Port of San Antonio were not designed to resist

the effects of seismic ground motions. However, according to stability analyses, it appe'ars that the wharf cranes can maintain

stability against earthquake ground motions equivalent to the horizontal seismic coeffi- cient of about 0.2. Therefore, it is concluded

that the wharf cranes may have sufficient stability and strength of its members. The stress and the instability from loading cargo may be severer than seismic ground motions. It can be presumed that the collapse of some wharf

cranes was caused mainly by the unfavorable

performance of wharf cranes foundations, such

as the quaywalls.

5. TECHNICAL COOPERATION

5 . 1 The First Mission

Immediately after the earthquake, the Chilean

government requested the technical cooperation

of the Japanese government. Major objectives

of the cooperation were:

i) to assess the structural reliability of

the damaged port facilities.

ii) to advise the Chilean government about the investigations necessary for designing and planning the restoration work.

393
The Japanese government organized the mission

of 6 experts (referred to the first mission, hereafter) and sent the delegation to Chile

from March 23 to April 5. The first mission

focused on the field investigation of the ports, preparing some analyses, and presenting

a report to the Chilean government. The report

included surcharge restriction of each berth

when it was necessary.

The first mission reported to the Japanese

government that the Port of Valparaiso should

be modernized, especially for the container

cargo, and that the restoration work should be planned taking the possible modernization into consideration.

5 . 2 The Second Mission

According to the report of the first mission,

the Japanese government decided to expand the technical cooperation to include port planning

for the modernization. To make necessai^

arrangement for further technical cooperation,

the second mission was organized with 7 members. They visited Chile from May 27 to June

11. The technical cooperation consisted of dispatching experts and studying the following

items: i) present situation of port facilities, ii) immediate restoration plan, iii) master plan, and iv) restoration and improve- ment plan. A part of the study would be per- formed by the Chilean engineers under guidance

of the Japanese study team.

5 . 3 Experts and Study Team

To implement the agreement, three experts were

sent to Chile for two months during the first

time and two experts for about one and half

months during the second time. The experts pro-vided advise and technology transfer to the Chilean government engineers. The team visited

Chile for about two months to collect necessary

information and to discuss the matter with the Chilean government staff. A portion of the study is being performed in Japan. The study

is expected to be complete in about 11 months.

6. REFERENCES

1) Saragoni, R. H. , P. S. Gonzales, and M. B. Fresard, Analisis de los Acelerograms del Terremoto del 3 de Marzo de 1985, SES I 4/1985, Departamento de dingenieria Civil, Universidad de Chile, 1985.

2) Lastrico, R. M. and Joaquin Monge, Engineering Aspects of the July 8, 1971 Earthquake in Central Chile, Proceedings

of the 5th World Conference on Earthquake

Engineering, Vol. 1, June 1973, pp. 468-

477.

3) Bureau of Ports and Harbours, Ministry of Transport, Technical Standards for Port and Harbour Facilities in Japan, The Japan Association for Ports and Harbours, 1980.

4) Noda, S. and T. Uwabe, Relation Between

Seismic Coefficient and Ground Acceleration for Gravity Ouaywall, Proceedings of the 6th World Conference on Earthquake

Engineering, 1977, pp. 1963-1968.

394
Origin date and time March 3, 1935, 19:46'54" (local time) Eticenter Latitude 33"U'2i" S Longi tude 72’02'2i" W Depth 1 5 km Magnitude (Richter scale) 7.7

Table 1 Earthqake event

location Components Ground condition Distance from

ports

NS E'N UD Univ. Federico

Santa Maria 0. 16i 0.179 0.125 Rock

1.4 km east

from Baron Pier

at Valparaiso Port

El Almendral 0.163 0.293 Reclaimed

land 0.8km south

from Baron Pier

at Valparaiso Port

LI i 11 eo 0.659 0.425 0.852 5km south from

San Antonio Port

Table 2 Maximum acceleration around

ports

Datcriptlon

BortH No. 1

1 to 3 4 1 B s 7 1 a 9 and 10 Dlftplacaieat of copelioo seawards

Hot rstarliable

lass than lOci SUcbt

4ci

170c» Nai. 74ci Has. 20ci Hai.

1 60ci Mai. Hetlieible 1

1 Horizontal slidinc

of quoTvall eleieota

lO-lSci Mai. bat 70ca at tb« coonectioo with bertb A 70ci Hai. Slllbt Heawr slidloc

partif at the offshore end Sliibt

1

Hearf

si iiiioc totally Tiltlof of quaytall

seawards Hocliciblo None Reoarkable Hetlicible 1- 2" Heelitibla

! "

Gap betvewo qaafwaU eleieots ObsarTed but siQital

10-20 ca Obseried but ainiaai Parts rewarkable30ci at the i end units I Retarlable

i "

Settleieot of apron

and bacIfof-pQrt yard

less tban 15ci

but 30ci parti;

at Bartb \ Haae; 3Bci Heav;

3Sci

20ci Hea*; I 40ci

Heavy , 30cw Hone Settleiont of railincs Slizbt at Berths

1 and 2, but las. lOci at bertb 3 lOcn Has. Hear; i 3-5C1 but IBci Hai. HeavT

j

Heai;

20ca 1 Nona Assastiant of of oral! daiacos

I m m '

1 " ! raj

l.e(enil;Eiteiit of daiaies 0; Ha daia(a I : Siiibt II ; Hediu* m: Sevare IV: Qreat not applicable

Table 3 Extent of damage to berths at Valparaiso port

395
Descriptiop

Berth Ho. j 1 and 2 | 3

1 4 6 and 7 Dlsplaceient of Collapsed 70ci-l00ci Heclitible Neelicible

1 Naiba occur Horizontal slidine

of ouavaall eieients Collapsed - - Sliebt -

Tiltinx Q( quaTvail

seataras Collapsed - - Sliebt

i -

Gap betveen

quareall eieients Collapsed - - Nil 1

1 -

Settleient of apron

and bacll-of-port card Groat Severe Sliebt 40ci at bacb- of-deck yard

1 SOci at the rear

end of deck Settleient of ratlines Great Heavy Sliebt Nil 1 Neelleible

Assessient of overall daiaeesIV IV

L«(eDd:£xtest of daiates 0: do dance I : Sllctit U: dediui UI : Severe IV; Great oot applicable

Table 4 Extent of damage to berths at San Antonio port

396
397
398

Fig.

6

Cross

section

of

Berths

6

and 8

Fig.

8

Boring

log

of

Berth

1

at

Valparaiso

port

at

San

Antonio

port
reclaimed

sand

399

Fig.

10

Cross

section

of

Berth 3

at

San

Antonio

port
400

Photo

2

Copeline

Deformation

of

Berth 5

Photo 4

Settlement

of

Back

Yard

of

Berth

at

Valparaiso

Port

at

San

Antonio

Port
401

Photo 6

Collapse

of

Wharf

Crane

of

Berths

and 2

at

San

Antonio

Port
OBSERVATION OF THE STRUCTURAL DAMAGE CAUSED BY THE 3 MARCH 1985

CHILE EARTHQUAKE

Summary of a Talk

by Mete A. Sozen^ Followiag by 10 seconds an event that appeared to

be a precursor shock of magnitude (body-wave)

5.2, a major earthquake of body-wave magnitude

6.9 and surface-wave magnitude 7.8 (NEIS) struck

40 km off the coast of San Antonio, Chile, at 7:47 p.m. local time on 3 March 1985. The resulting strong ground motion affected a large

area which more than half the population of Chile

lives. The affected zone includes coastal towns

of Valparaiso and Vina del Mar. The total direct

and indirect damage was estimated to approach two billion dollars (U.S.). Because of the occurrence of major events in that area with a return period of almost a century and because of the expectation of another event between the years 1980 and 2000, an extensive strong-motion instrument network had been

installed in the region as a joint project of the Organization of American States and the University of Chile. This network yielded 31 strong- motion recordings with horizontal accelerations

ranging up to 0.67g. This collection of strong- motion data is certainly the most important

aspect of the 3 March 1985 Chile earthquake from

the viewpoint of structural engineering because

structural damage was, on the whole, light for recent construction. The number of casualties attributed to the earthquake was 176 which also indicates that structural performance was generally satisfactory, considering the number of population centers affected. Had such a con- vincing collection of strong-motion data not been

available, the event might have been dismissed as one corresponding to relatively weak ground

motion, despite its magnitude. As in the past,

the few failures would have been attributed to flaws in design or construction. The successes

would have been ignored.

An important feature of the strong-motion data obtained for this event caused by a subduction

fault is its spatial distribution. Figure 1 shows the scatter in the variation of the maxi- mum horizontal acceleration with epicentral

distance. The data in Figure 1 have been plotted

to linear x- and y- scales. Figure 2 shows the same data plotted logarithmically. Even though

the logarithmic representation appears to Imply

the existence of a trend, it does not change the facts which require serious study in view of the practice of using the epicentral distance as the dominant factor in estimating ground-motion

intensity.

The acceleration response spectra calculated

( B = 0.02) for the 1940 El Centro N record

(Amax = 0.33g) and the 1985 Vina del Mar S20W

record (A^gx = 0.36g) are compared in Figure 3. (The spectrum for Vina del Mar was based on values reported by R. Sarragoni.) In terms of instrumental indication, the two events would be expected to have comparable damage potential.

But the overall record of the large inventory of buildings in Vina del Mar was quite good. Given

that the design-code requirements for lateral strength were comparable to those used in zones

of high seismicity elsewhere in the world and that the detail requirements for the reinforced

concrete buildings were not demanding, it follows

that detailed studies of the reasons for the satisfactory performance of the Vina del Mar buildings may have a very strong impact on building construction for earthquake resistance.

402

^Department of Civil Engineering, University of Illinois, Urbana, IL
Measured

Maximum

Acceleration

VARIATION OF MAX. ACCELERATION WITH DISTANCE

O' 1 .00

0.75

0.50

0.25

0.00

0 1 00 200 300

Distance from Epicentral Region,

400

km

Figure 1

403
Measured

Acceleration/g

Logarithmic Representation

404
Response

Acceleration,

Vina d Mar 1985 and El Centro 1940 (Damp. Fact. = .02)

Period, sec.

Figure 3

405

Papers Not Presented Orally

But Included In The Proceedings
. i ;Ub9b4’!:i>fi| lijS
SPATIAL STRUCTURE OF THE WIND IN THE MARINE BOUNDARY LAYER

BY Keikichi Naito*, Tatsuo Hanafusa**, and Tokunosuke Fujitani**

ABSTRACT

This study investigated the temporal and

spatial structures of the wind in the marine

boundary layer. For this purpose the array of instrumented towers is located at the seaside.

The results of preliminary analysis of low-wind

speed data are presented. The eddy is hori- zontally isotropic in unstable condition, but

is elongated along the mean wind direction in stable condition. In unstable condition the lateral coherence function of longitudinal

velocity component followed exponential decay

formula, but that of lateral component followed

approximately the theoretical formula derived

from the theory of isotropic turbulence.

1. INTRODUCTION

There have been many studies of turbulent

characteristics of the wind in the marine

boundary layer. However, there remains insuf- ficient knowledge of the spatial structure of

the offshore wind, because wind measurements at various points on the sea surface are difficult

to perform. In the present study, an array of instrumented towers is located at the seaside.

The preliminary observation begun in 1984, and

overall data collection started in March of

1985. High winds have not been observed, but

the results of preliminary analysis of low-wind

speed data are presented.

2. OBSERVATIONAL SITE AND INSTRUMENTATION

The observational site is located at the seacoast of the Pacific Ocean, about 100 km

east of Tokyo. The shoreline is straight and

the topography is flat and open. One 36 m and

three 20 m towers form the observational array

which is parallel to the beach about 70 m down- wind from the shoreline. The dunes of 2 m high

are located about 30 m apart from the array as shown in Fig. 1. The separation distances of

20, 40 and 80 m are used. Three 3 -cup anemometers are provided at 10, 20 and 35 m for vertical wind profile measurement and two resistance

thermometers are installed at 10 and 35 m. The 2-dimensional sonic anemometer at 35 m high can measure wind speed and direction, and these data are fed into the automatic sequence con- troller for unmanned operation of the data

* Meteorological Satellite Research Division. **Meteorological Research Institute, Japan

Meterological Agency, Tsukuba, Ibaraki 305, Japan.

acquisition system. The 3 -dimensional sonic anemometer- thermometers are provided at the

20 m level of each tower for measuring

fluctuations of wind velocity components and temperature

.

The preliminary observation was executed by

using a 36 m and two 20 m towers.

3.

AERODYNAMIC CHARACTERISTICS OF THE SITE

In this study, the observation was performed

near the seacoast, thus it is necessary to evaluate the effect of the modification of the wind field induced by the change of surface

roughness at the land- sea interface. Following

the equations of Jackson (1) and Wood (2), it

is possible to evaluate the height of the internal boundary layer developed from the land- sea interface. The roughness height over

a sandy beach is assumed to be 0.1 cm. The estimated value of the internal boundary layer height is about 2 m or 3 m. This value is small enough to compared it to the lowest observational height of 10 m. To ascertain the aerodynamic characteristics of

the site, vertical wind profiles were obtained.

The wind speed ratios averaged for total sampling duration are shown in Fig. 2. The profile follows the logarithmic law; the estimated roughness length is about 0.08 mm, which is compatible with other results obtained

over open ocean. Result suggests that the observed wind, at this site, has the characteristics of offshore wind.

The data discussed in this study were obtained

under stable and unstable conditions. A sam- pling rate of 50 or 100 msec was used for mea- surements and in the analysis raw data were averaged for every 1 second. The data were divided into sub-data blocks for 10 minutes,

and turbulence statistics were computed for each sub-data block.

To investigate the spatial extent of the eddy,

the integral scales of each velocity component

were evaluated from correlation functions. The integral scale of longitudinal component (u) was evaluated from autocorrelation curve by

assuming Taylor's hypothesis. In unstable con- dition, the block- to-block and tower- to- tower variabilities of the integral scale are not so large. On the other hand, in stable condition

both variabilities are somewhat larger. The mean values of integral scale of u component

409
are 126 m in unstable and 81 m in stable

conditions, respectively. The lateral extent

of the eddy was evaluated from the lateral

spatial correlation of lateral velocity

component(v) . The cross correlation analysis

shows that the correlation of v component is larger than that of u component and also the correlation of vertical velocity component (w)

is negligibly small with the separation

distance longer than 20 m. The correlation

peak of u component shifts progressively as increasing separation distances. In stable

condition, the correlation of v component is also larger than that of u component and the correlation of u component with the separation

distance longer than 20 m are very small. Figure 3 shows the variations of the cross correlation in lateral direction. In the preliminary observation, the separation

distances are insufficient to carry out the numerical investigation. Therefore, in the present study, the exponential decay assumption

was applied to these data. Estimated integral

scales of each velocity component for longitu- dinal and lateral directions are given in Table

1. The main features of this table shows that

in unstable condition the scale of u component

was almost the same as that of v component.

This fact suggests that the eddy seemed to be horizontally isotropic in unstable condition. Meanwhile, in stable condition, the scale of u component was about 3 times larger than the scale of V component, that is, the eddy was elongated along the mean wind direction.

Figure 4 shows the lateral coherence of u and v

components in unstable and stable conditions.

In unstable condition the coherence of u component can be well expressed by the exponential decay function, and the broken line is corresponding to the curve with the decay parameter k-8 . On the other hand, the coherence of V component is determined by a slightly

different formula. The similar shape to this curve can be derived theoretically from the homogeneous and isotropic turbulence

(Kristensen and Jensen (3)). As mentioned

before, the eddy in unstable condition was horizontally isotropic, thus, the present

result obtained here is probably reasonable.

In stable condition, lateral coherences also

are well expressed by the exponential decay

formula.

To investigate the spatial structure of the offshore wind, the instrumented tower array is provided near the seacoast and preliminary

observation was executed. The evaluated

integral scales of turbulence suggest that the eddy is horizontally isotropic in unstable

condition and is elongated along the mean wind

direction in stable condition. The lateral coherence functions except v component in unstable condition can be described by

exponential decay formula. The coherence of v component in unstable condition can be described by different formula approximately

similar to the theoretical result of homogeneous and isotropic turbulence. The authors executed the overall measurements using good collected data. They will present the results

of their detailed analysis in the near future.

6. REFERENCES

1) Jackson, N. A., 1976, The Propagation of Modified Flow Downstream of a Change in Roughness, Q. J. Roy. Met. Soc., 102, 924-

933.

2) Wood, D. H. , 1982, Internal Boundary Layer Growth Following a Step Change in Surface

Roughness, Boundary Layer Met., 22, 241-

244.

3) Kristensen, L. and N. 0. Jensen, 1979, Lateral Coherence in Isotropic Turbulence

and in the Natural Wind, Boundary Layer

Met., 17, 353-373.

Table 1 Integral scales for each velocity component.

Stability Direction U V w Unstable Longitudinal 126m 218m 37m Lateral 52m 134m - Stable Longitudinal 81m 77m 26m Lateral 17m 28m -

410
i Shore

, 36m Tower

20m Tower

Fig. 1 View of tower array.

Fig. 2 Vertical wind profile.

41
msc (unstable) 8^^ 06 C (stable)

dib 5 u

lO

(/) ou

u

w

0

^ 1 ^ 1 Ly 0 40 80 m

ly (0) (66) 0

(0)

w

80 m

(79) -.3

Fig. 3 Spatial crosscorrelation for lateral direction.

unstable(S40CC)

Ucofi|jy(nly/0)-exo(-iuily/0l k-8

y ni /u

.1 .2 _

.3 .4 .5 nIy/U

Fig. 4 Lateral coherence of u and v components.

412
LOADING TEST ON THE CONNECTION

OF CONCRETE FILLED STEEL TUBES

BY Hirohiko Tada , Hiroshi Shinohara , Koichi Minosaku , Akira Takizawa , and Jun Hikino

ABSTRACT

Concrete filled steel tubes have good

load-resisting performance as a com- posite structure composed of steel and

concrete. However, it is important to examine the load-resisting performance

of the connection using concrete

filled steel tubes to reflect the

actual design of bridge piers in urban

expressways and monorails.

Five different models of the connection were made; static loading

tests were performed to investigate

the difference of the load-resisting

behavior with respect to the existence

of filled concrete, studs and

diaphrams, and understand the region

of filled concrete. Loading test results showed little

difference in the capacity and the performance of deflection of each

speciment; on the whole the specimens

with filled concrete show good

loading-resisting peformance.

1. INTRODUCTION

As described in previous reports ^)-

, concrete filled steel tubes

(hereafter called "CFST") have large

capacity and good ductility as a composite structure composed of steel

and concrete. The application of CFST

to bridge piers in urban expressways

and monorails require large capacity

and aseismatic performance. However,

the peformance of CFST is confirmed

for simple members such as a beam and

a column. There is no complete data

on its peformance for applying to actual structures.

There is concern that the collapse

of the connection leads to the collapse of a structure in cases when the CFST is not adequately

designed.

In general, conditions required for the connections required for the connection are:

(1) withstand collapse before bridge piers show original strength and ductility.

(2) possess ample strength and rigidity to be satisfied with the assumption in the calculation that the part of connection is a rigid

point.

Against these requirements, there

are many studies on connections

of steel piers, and practical

design methods, presented by

Okumura et al. . There are few studies about the connection when using CFST as a column; the design method and structual details are not filled

concrete, its connection needs to have higher load- resisting performance

than a steel column. The author intented to determine the desirable CFST connection in this loading test. Five models of the connection in rigid frame piers were

made, and static loading tests were performed to investigate the difference of the load-resisting

behavior with respect to filled

concrete, studs and diaphrams and the region of filled concrete. Then, the estimation of the specimens rigidity

by two dimensional finite element method and the comparison of test results and theoretical results by the calculation program for beding

capacities under changeable axial

force were performed.

I) Director of Structure and Bridge Department, PWRI

II) Chief of Bridge Division, Structure and Bridge

Department, PWRI

III) Research Engr. of Bridge

Division, Structure and Bridge Department, PWRI

IV) Research Asst, of Bridge

Division, Structure and Bridge Department, PWRI

413
2 . OUTLINE OF TEST METHOD

2 . 1 Specimens

As described in the introduction, five types of models of the connection in rigid frame typed piers were made to determine the desirable type of connection CFST. The dimensions and

the specifications for each specimen

are indicated in Table-1 and Figo-1.

Ready-made steel tubes with thin walls

were used as a column. Beams were made of steel plates.

Stiffened plates are attached to eliminate local deflection from concentrated forces near the applied

point of force. Double rigid steel plates are arranged at the lower edge

so the specimen can be fixed in the bed of the testing machine.

In the design of specimens, the beam

is designed to fail only after the collapse of its column connection.

This was done to understand the difference of the load-resisting

behavior for each specimen with

the following features (see. Fig. -2)

Type E This specimen is similar to Type B, but studs

are arranged near the joint

between the beam and column. The stud is expected to prevent sliding in the boundary

between steel and concrete and the separation of concrete from steel due to the difference of Poisson's ratio in steel and concrete. The Concrete used 'has specifications as shown in Fig. -2 so compressive strength of 240kg/cm2 can be obtained.

Average compressive strength of three cylinders is (10 cm in length and 20cm in diameter) , at seven

days is 251kg/cm^.

2 . 2 Loading method

Loading tests are performed using the 3000 ton large universal structual testing machine installed in PWRI's structure engineering laboratory. Each specimen is fixed in the testing

machine bed by bolts not subjected to eccentric force. Photo-1 shows the specimen.

Type A This is the standard specimen without filled concrete

in the beam and column. A diaphram is attached to prevent

local buckling in the inside of

a colunm adjacent to the lower

flange of a beam. Type B This specimen has filled concrete in the column in part of the beam; it does not have a stud or a diaphram. A thin steel plate is fixed in place of the form to fill into concrete in the designated part

of beam. Type C This specimen has filled concrete in the column to determine the need for concrete

in a beam. Type D This specimen is similar to Type B, but a solid

diaphram is arranged in a beam. This diaphram is expected to prevent the sliding in the boundary between steel and concrete and the volume

expansion of concrete in the plastic region.

A static load is applied to the top

plate of the specimen until specimen's

horizontal displacement reaches the allowable stroke of the testing

machine controlling the load in the region of low load and the displacement in the region of high

load. The upper edge is possible to move freely in the horizontal

direction and the rotation by the above equipment for loading.

2.3 Strain gauges and displacement

indicators are set as shown in Fig.-

3(a) and (b) . Strains on the surface

of steel and displacements at prescribed points are measured at fixed loads. Particularly, three directional guages are attached on the surface of steel near the joint

between a column and a beam so detailed distribution of the stress

can be measured. Also, the change of deflectional angle at the joint and

the lower edge is indirectly measured

by setting the rolling typed

displacement indicators. The separation of mill scale and the out- of-plane deflection are observed

during the tests.

414
3. TEST RESULTS AND DISCUSSION 3 . 2 Capacity

3 . 1 General

Type B to Type E with filled concrete

in the columns show large capacity and sufficient ductility in comparison

with Type A without concrete.

Figs. 4(a) to (e) show the situation after loading for each specimen (ref. Photo -2 and 3)

.

Type A indicates large out-ofplane deflection near the joint

in a column at almost maximum

load, and its deflection grows

in proportion to the increment of load. Simultaneously, the separation of mill scale is seen only at lower flange.

Type B shows the separation of mill scale near the joint in a column at a load of 170 ton. Thereafter, the out-of-plane

deflection near the joint in the lower flange of a beam is observed.

It is recognized that the performance of Type C to Type E

is approximately same as that

of Type B. However, Type C indicates the mode of local buckling toward the inside of

a beam near the joint. This

probably results because this

specimen type does not have

filled concrete in the beam.

Also, the out-of-plane

deflection near the joint in beams of Type D and Type E is slight compared with that

of Type B. This may be caused

by the stud and diaphram.

Filled concrete prevents the out-of-plane deflection toward

the inside of steel and the arrangement of studs controls

the out-of-plane deflection

toward the outside of steel.

Table-3 shows theoretical results

and test results related to three kinds of load. The following are assumed in the calculation of capacities.

(1) No slip occurs in the boundary

between steel and concrete. That is, the distribution of longitudinal

strains at a section follows the assumption of the plane preservation

(bending cracking load and steel yielding load)

.

(2) Neither shear failure nor local buckling occurs.

(3) The tensile strength of concrete

is negligible in the calculation of steel yielding load and load corresponding to the full-plastic

resisting moment.

(4) The stress-strain relationships of steel and concrete follows the relationship as shown in Fig. -5 (a),

(b) respectively.

Based on these assumptions, the section is divided into small elements

as shown in Fig. -6, and numerical

analysis is performed using Newton- Raphson method with two parameters,

which are the position of neutral axis

e and the local P. In the analysis, the initial values of the veritical load and the position of neutral axis, Pq, e°, are obtained by

the elastic theory, and in turn are corrected in accordance with equation-

(1 to equation- (4) , so that the equilibrium of force can be satisfied

and strains of upper or lower fiber at

a section become prescribed become prescribed values.

415
-Equilibrium of force-

(1) To axial force

^ Fn=N+ { I (ask- Ask+Cfck-Ack) - Z ( Cfsj- Asj+ (Jcj- Acj) } = 0

k=l j=l

(2) To bending Dooent Fa-M- { Z (CJsk-Ask* Zsk+<?ck-Ack" Zsk)

k=l n — Z (CJsj" Asj* Zsj+ (Jcj* Acj* Zsj) ) = 0

j=l

in vhich, N : Axial force due to external force (kg) Coluan : N = P • cos d Beaa : N = P • sin 9 M ; Bending moaent due to exterai force (kg.ca) Column : M= P • cos 9 x (11.2+ e) + P • sin 9 X168

Beaa : M= P • cos 9 X 109.5— P • sin 9 X e

• ( 1 )

Ap

Ae

Fin(p,e)’Gn,e — Fn(p,e)‘Gm,e Ga»P'Gn,e — Gn,p-Gn,e

Fn(p,e)'Gm,p— FB(p,e)*Gn,p

Giii,p*Gn,e — Gn,p*Gm,e

( 2 )

Gn,p

Gffl,p

Fn(p+Apo,e) — Fn(p,e) Apo F m(p+ Apo.e) — F !j(p,e) Apo

Gn.e

Gm,e

Fn(p,e+ Aeo) — Fn(p,e)

^ Aeo

F m(p,e+ Aeo) — F m(p,e) Aeo /

•

• (3)

e Aeo = ——

1000

(4)

It is assumed that the capacity

of all specimens are determined

by the capacity of a column as

in the design of specimens from Table-3. This is also realized

from the fact that the out-ofplane deflection is large near

the joint. However, as long as the Type C without filled concrete in a beam is concerned,

the influence of the out-of-plane

deflection in a beam upon the capacity is realized because the capacity of this specimen is larger than the load corresponding the full-plastic resisting

moment. Its capacity is slightly smaller than that of Type B, D, E. Type A indicates the incre- ment of out-of-plane deflection

from near the steel yielding

load, and its capacity is considerably smaller than the

load corresponding to the full- plastic resisting moment. Filled concrete inside a column or a beam has sufficient resistance

against the load and holds the increase of strength after steel yielding. This occurs because

filled concrete prevents the development of out-of-plane

deflection of steel. Compared calculated values with

measured values for the steel yielding

load, a measured values are smaller

than calculated values in all specimens. And especially, measured

values of Type A and Type C related to

a column are comparatively smaller

than calculated values. Therefore,

the influence of shear lag is needed

in the case of connections as Types B,

D, and E, and certainly for Types A and C.

416
3 . 3 Rigidity and ductility

Fig. 7 to 10 show the relationships

between the yertical load and yertical

displacement, horizontal displacement,

the ratio of yertical displacement to horizontal displacement, and the deflect ional angle at the joint

respectively. Furthermore, the

results of linear analysis by the

finite element method, in which

specimens are replaced to the two dimensional plate model as shown in

Fig. 11, are indicated in Fig. 7 & 8, and Fig. 10. S-line shows load- displacment relationships provided

that flexual rigidity of only steel is taken into account, and S+C-line

indicates both concrete and steel. The ratio of Young's modulus in steel

and concrete is considered in the evaluation for rigidity of filled

concrete. The plate thickness and the

plate width of analytical model are determined so that a fiber strain

in specimens and the analytical model

can be coincide. That is, area of cross-section A and section modulus W

in the following equation are made to

be equal in specimens and the analytical model.

in which: M;

Fig. -3 indicated previously that bending cracks occur; and the relationship between the load and the displacement becomes uni inear

in Type B to Type E. However,

it is found from Fig. -7 and 8 that the relationship between

the load and the displacement

including that of Type A are almost linear up to the steel yielding load, the thereafter

the displacement becomes larger.

Compared the slope of the straight line of Type A with

that of Type B to Type E before

the steel yielding; it is realized that the rigidity of specimens becomes large by the effect of filled concrete. For instance, the displacement of Type B to Type is almost 50% of that of Type A at 50 ton load. Furthermore, the ductility of Type B to Type E is improved by

filled concrete compared with

that of Type A because the load decreases in Type A and not observed in Types B to E.

( 3 )

lonent due to external force

Axial force due to external force

The relationship between the

load and the ratio of horizontal

displacement to vertical

displacement shown in Fig. -9

is dispersed by the setting

error and measuring error in the range of small load. Thereafter, its relation

continues to be an almost

constant value; it increases

near the maximum load. This

means that the deflection of

a column abruptly increases

near the maximum load, and the capacity of specimens is determined by the capacity of the column.

It is determined that both

filled concrete and steel resist

the load in theoretical

calculation, because measured

values of Types B to E in Fig. -7 and Fig. -8 lie on S+c-line approximately. Measured values of horizontal displacement

of Type A in Fig. -8 lie on almost Sline, but measured values of vertical

displacment of Type A in Fig. -7 is considerably larger than theoretical

values. This reason is Type A is affected severely by the shear lag and

the out-of-plane deflection compared

with other specimens.

417
3 . 4 Characteristics of strain

Fig. -12 and -13 show the relationship

between the load and the tensile

strain at positions indicated. Fig.-

14(a) to (e) show the distribution of longitudional strains in a column at representative loads. First, it is shown from Figs. -12 to -13 that type

a beam yields after column yielding

near a load of 60 ton; there is little

increase of strain in a beam after

column yielding. This means that only

the column goes toward collapse.

Types B to E suggest almost the same relationships between the load and the

strain. The strain in a beam

increases after column yielding

because the column does not collapse

(due to the filled concrete) . It is determined from Fig. -14 that the influence of shear lag in Type C and

Type A is larger than that of other

specimens.

4 . CONCLUSION

The following results were obtained

from these loading tests.

(1) The influence of shear lag can be

reduced.

(2) The out-of-plane deflection can

be controlled, and the strength in the plastic region is expected.

Large carrying capacity greater than

the full-plastic resisting moment and good ductility resulted.

The ductility and characteristics of strain. Type B, Type D and Type E which have filled concrete in the column and the beam are desirable

types of connection when concrete

filled steel tubes are used as a column.

It was determiend that the rigidity of concrete filled

steel tubes in the elastic region

can be evaluated as a composite

structure provided that the ratio

of Young's modulus in concrete

and steel is taken into

consideration.

REFERENCES

1) Narita, N. , Saeki, S., Kanai,

M. , Oshio T., "Behavior of Concrete Filled Steel Tubes

(Part 1; Compression Members)",

Technical Memorandum of PWRI

No. 1771, February, 1982.

2) Narita, N. , Saeki, S. Kanai,

M. , "Behavior of Concrete

Filled Steel Tubes (Part 2; Bending Members)", the 15th Joint Meeting of U.S. -Japan

Panel on Wind and Seismic

Effects, UJNR, May, 1983.

3) Saeki, S., Minosaku, K. , Takizawa, A., "Behavior of Concrete Filled Steel Tubes

(Part 3; Beam-Column Members)", the 16th Joint Meeting of

U.S. -Japan Panel on Wind and Seismic Effects, UJNR,

May, 1984.

4) Tada, H. , Shinohara, H.

,

Minosaku, K. , Takizawa, A., Hikino, J. J., "A Study on Behavior of Concrete Filled

Steel Tubes", Proc. of the 2nd

J Task Committee, September,

1985.

5) Okumura, T. , Ishizawa, N. , "The Design of Knee Joints for Rigid Steel Frames with Thin

Walled Section", Transactions

of the Japan Society of Civil

Engineers, No. 153, May, 1968.

418
Table. 1 Specifications for Specimens Type of specimens A B C D E

Column

Diameter (mm) 609.6 609.6 609.6 609.6 609.6 Thickness (mm) 9.5 9.5 9.5 9.5 9.5 Designation STK41 STK41 STK41 STK41 STK41 Yield strength

(kg/al) 3,500 3,500 3,500 3,500 3,500 Breaking strength

(kg/cirf) 4,800 4,800 4,800 4,800 4,800

Beam

Width (mm) 450 450 450 450 450 Height (mm) 450 450 450 450 450 Thickness (mm) 19 19 19 19 19 Designation SS41 SS41 SS41 SS41 SS41 Yield strength

(kg/caf) 3,000 3,000 3,000 3,000 3,000 Breaking strength

(kg/ai) 4,500 4,500 4,500 4,500 4,500

Concrete

Column without with with with with Beam without with without with with

Concrete strength (kg/caJ) - 251 251 251 251

Stud without without without without

with 9.5« X80

Oiaphram without without without

with t=16mm

(SS41)

without

Remarks

effect of diaphram

effect of stud

Table. 2 Specification for concrete

Nominal

strength

Unit

(rn)

Cement C

(kg)

Water W

(kg) W/C

(%)

Fine aggregate

(kg)

Coarse aggregate

(kg)

Slump

(cm)

Air content (%)

240 1.0 251 148 59.0 839 1,119 8.0 4.0

419
Section A - A

Section B - B

Figol Configuration of Speciaen

420
Fig. 2 Detail of Connection Part

421
422

Three-Directional

Guage
423

Siluation

of

Speciaens

after

Loading
1) C4ic«i«e«4 V«iu« of MaiiMfl Lo«d indicates Load corrsaaondinc to eha fuU'sUacie rasistinc 2) E acA Load »Ad«s vaiuas o{ Saceion A^A in a cdLusa and S aeciaa 9*3 in a ^a.

Strain i :

(a) Steel (b) Concrete

Fig. 5 Stress— Strain Curve

Fig.G Element Division of Section for the Strength Calculation

424
425

....I

..nil

Horizontal

Displace.ent

Kig.lO

Relation

of

Load

and

Deflectional

Angl,
426

Deaa
427

I'

ig.l4

Ui

blr

ibul

ion

of

Lontiiludioiiiil

Strains

ina

Coluan
Photo .

1

Setting Situation of Specimen

Photo. 2 Situation after Loading

(Type-A) Photo. 3 Situation after Loading

(Type-0)

428
DEFORMATION CAPACITY OF STEEL TRUSS GIRDER

BY Isao Nishiyamai , Tatsuo Murota^ , Nagahide Kani^ , and Shin-ichi Ina^

ABSTRACT

I The aim of this research is to make clear the

I

relation between the deformation capacity of

I steel truss girder and the slenderness ratios

of the individual members or the failure mode

of the steel truss girder. Six steel truss

i: girders, composed of double angle section

chords and braces, were loaded monotonically

.

Five girders of H-section chords, arranged by

their weak axis in the plane of the girders,

and circular tube braces were tested by cyclic

loading.

The following results were obtained: (1) large deformation capacity can be anticipated, in a

I steel truss girder, if the slenderness ratio of individual members are small. (2) Steel truss girders of chord failure type should be

I restrained from its lateral buckling to ensure

I deformation capacity. (3) Steel truss girders

of brace buckling failure type have little deformation capacity without slippages of joint

bolts. (4) Overall deformation characteristics

of steel truss girders can be evaluated as the sum of the deformation characteristics of the individual members which construct the girder.

(5) Inelastic behavior of steel truss girders

are largely affected by the plastic deformation

of a few individual members near the end of the girder.

According to the current seismic design

standard in Japan (Ref. 1), both strength and deformation capacity are required for steel building structures. For full web members,

many experimental investigations were conducted

and the deformation capacity and the strength

was well documented. For open web members such

as trusses, however, research is needed to evaluate the deformation capacity. To obtain

the fundamental data on the deformation

capacity of steel trusses, bending tests were performed in two test series.

In the first, monotonic loading tests of six

steel truss girders with double angle section

chords and braces were performed. Here, the ^Research Engineer, Structure Engineering

Department, Building Research Institute,

Tsukuba, Japan. ^Director of Structure Engineering Department,

BRI. ^Member of Structure Tech. Special Committee, Japan Federation of Architect Office

Association, Tokyo, Japan.

rough consideration was done on the relation between the deformation capacity and the slenderness ratios of the chords and braces of

the truss girders-. In the second series, cyclic loading tests were performed on five steel truss girders with H- section chords arranged by their weak axis in

the plane of the girders and circular tube braces. The interaction of the inelastic behavior of the girders and the sectional

shapes, slenderness ratios of the individual

members used, or failure modes of the girders

was considered. Moreover, each axial deformation of the individual member of the truss girder was measured. From the measured axial deformations, the overall deformation of the truss girder was estimated. The estimated

deformation was compared with the overall deformation directly measured in the test.

2. TEST SPECIMEN

2 . 1 Series 1 Test

The representative shape of the six test specimens and its joint detail are shown in

Fig. 1. Both top and bottom chords were made

of a double angle section ( 2L- 65x65x8 )

.

Diagonal braces also were made of a double

angle section (2L-50x50x4) . The ends of the braces were frictionally connected to the chords through the gusset plates by three high- strength-bolts. Only the depth and the shear

span length of the truss girders were varied

but the sections used and joint details were common in these specimens. See Figure 1 for and chords and braces. CU- , CD-, and B- are

top chord, bottom chord, and diagonal brace

respectively. The number, in these indexes, is counted from the left-hand side end of each

girder. The dimensions of the test specimens

are summarized in Table 1. The mechanical

properties of steel are listed in Table 2.

2 . 2 Series 2 Test

The representative shape of the five test specimens is shoim in Fig. 2. Both top and bottom chords were made of H-section members

arranged by their weak axis to coincide with

Che plane of the truss girder. The braces were circular tubes, and the angle between chord and

brace was 60 degree. Chord- to-brace joints

were of fillet welding using gusset plates, which were arranged with a right angle to the

429
plane of the strong axis of the truss girder

and inserted into the circular tube braces by

cutting slits. Failure modes and slenderness

ratios of the members, which seemed to fail, varied for each specimen. The dimensions of

the test specimens are listed in Table 3. The mechanical properties of steel are listed in Table 2.

3. OUTLINE OF LOADING AND MEASUREMENTS

3 . 1 Series 1 Test

The test set-up and the measuring

instrumentations are schematically shown in

Fig. 1. Both ends of the test specimen were pin-roller supported and monotonic vertical

force was applied concentrically just on the

top of the vertical stiffener to induce triangularly distributed bending moment. Lateral deflections were fully restrained at both ends and intermediate loading position of

the test specimen. The rigid part of the specimen shown in Fig. 1 was much higher in rigidity than the testing part, so the rotation

about the weak axis of the truss girder did not

arise at the position of loading. Measurements were conducted so the deformation

of the specimen, as a cantilever beam, could be computed from measured data.

3.2 Series 2 Test

The test set-up is shown in Fig. 3. The test specimen is a cantilever type girder; one end

fixed to the reaction wall and the other end subjected to vertical force by an actuator

through T-shaped loading-rig. The out-of-plane

truss deformation were laterally restrained at

the *-marked locations in Fig. 3. Three loading cycles were started from negative

loading. They were carried out with the final collapse under the last positive loading, where

the upward loading is considered positive

loading in Fig. 3. Vertical and rotational

deformations at both ends of the testing part

of the specimens and axial deformations of CU1 and -2, CD-I and -2, and B-1, -2 and -3 were measured as shown in Fig. 4. Axial strains of the members near the end subjected to maximum stresses were measured in both series of tests, but they are not referred

to in this paper.

4 . 1 Series 1 Test Maximum strength and failure mode etc. of each specimen are summarized in Table 4. The applied load (shear in the testing part) vs.

displacement relationships of the specimens are shown in Fig. 5 to Fig. 10. The yield strength of the specimen T-66-21 was about 14 tonf; after the yielding load carrying

capacity increased to 16.4 tonf by strain hard- ening until it was interrupted by the buckling

of B-1 brace (Fig. 5). However, the deterioration of the load carrying capacity was rela- tively small even after the buckling of the brace. The severe reduction of the strength of

the specimen was observed after the lateral buckling of the girder. As for the maximum

strength, the test result determined by brace buckling is far smaller than that calculated as tabulated in Table 4. The high stress concen- tration and the large eccentricity of diagonal

brace is not considered in the calculation.

In specimen T-66-40 (Fig. 6) and in specimen T- 99-40 (Fig. 7), brace yielding began at about

15 tonf, and the deformation increased with

slippages of the high-strength-bolts connecting

the braces to the chords. The broken lines in

Figs. 6 and 7 show the reduction of load carrying capacity by the slippages. In the specimen T-66-40, the load carrying capacity

increased to 16.2 tonf until disturbed by the buckling of the B-1 brace. After the reduction

of the load carrying capacity, the trapezoid

frame, which consists of CU-1, B-2, CD-I and

the vertical stiffener at the loading position, showed relatively stable restoring force characteristics. The stable strength was about

12.0 tonf. The specimen T-132-60 failed by brace buckling

as did specimens T-66-40 and T-99-40, but no slippage was observed, thus its deformation

capacity was poor. Both ends of the buckled braces rotated about

each middle bolt of the end-connections, so the effective buckling length of the braces could

be considered to be the length between both

middle bolts. Lateral buckling occurred at 12.9 tonf in the specimen T-132-40 (Fig. 8) and at 11.1 tonf in

the specimen T-165-60 (Fig. 10). After the lateral buckling, the load carrying capacity

decreased in both specimens. These specimens,

in which lateral buckling occurred in the early

stage of loading, showed little deformation

capacity.

The elastic energy absorption capacity defined

by Eo-Qy6y/2 and the overall energy absorption

capacity Ek were estimated from the load vs. displacement relationships, and the specific

value Ds in the Japanese seismic design

standard (Ref. 1) was calculated from Eq. (1). According to the current seismic design

standard in Japan, the value Ds of most ductile

steel structure is defined to be 0.25 and it increases as the structure becomes less

430
ductile. The estimated values are listed in truss girder can be predicted from the Table 5, where Qy is deformation of each individual member.

Ds - Ek/Eo

( 1 )

the yield strength, 5y is the yield

displacement estimated by the division of yield

strength by elastic stiffness, and Ek is the energy absorbed until the load carrying

capacity of the truss decreases to k% of yield

strength. The calculated Ds-values from Ek with k“100% widely varied from 0.26 to 0.58. These calculated results show that a truss girder can have large deformation capacity if individual members are selected with small slenderness ratio.

4.2 Series 2 Test

These test results are summarized in Table 6. The load vs. displacement relationships of the specimens are shown in Figs. 11 to 15. In these figures, the displacement is measured

vertically at the right-hand side end of the specimen and the upward displacement is positive. Both specimens A and E showed plastic flows

(flat portions) in the load vs. deformation

relationships under the positive loading,

produced by the tensile yielding of the CD-I prior to the reduction in strength by the buckling of the CU-1. The slenderness ratio,

A, of the CU-1 in the specimen E was 34, and it was much larger than that of the specimen A, so

the ratio of the buckling strength of the CU-1

to the tensile yield strength of the CD-I was relatively smaller in the specimen E than in specimen A. Therefore, the flat portion of the load vs. deformation relationship was shorter

in specimen E than in specimen A. Under the negative loading, the load carrying capacity

decreased just after buckling of CD-I (A>10)

.

The deterioration slope in specimen E was much

steeper than in specimen A, because the slenderness of the buckling member in specimen

E is larger than that in specimen A. Axial force vs. axial deformation relationships

of the principal members of the truss girder

are shown in Figs. 16a to 16f for specimen A, where the axial force was calculated based on

the assumption that was the determine truss.

The CU-1 and CD-I mainly shows the inelastic

behavior; the others remained elastic. Measured axial displacements for the members

and calculated ones for the other members are used to estimate the overall girder deformation. The estimated overall deformation is shown in Fig. 11 as the dotted line. From the comparison of the solid line (test result) and

the dotted line (evaluated results), it can be concluded that the overall deformation of the

The specimens B, C and D reached their maximum

strength by brace buckling in positive and negative loadings showing non-ductile hysteresis loop. Especially in specimen B, a very

brittle type of snap through buckling (Ref. 2) occurred. After the reduction of the strength

by such brace buckling, these specimens showed

rather stable hysteresis loops, which resulted

from the contribution of the frame activity

prescribed on the trapezoid frame action in the specimen T-66-40 in the series 1 test. The practical yield strength of this stable loop can be explained by a virtual work method

considering the positions of the yield hinges

as shown in Fig. 17. From these test results in general, truss girders of the type of brace buckling failure

is non-ductile. Of course, the specimens T-66-

40 and T-99-40 showed relatively large deformation capacity in spite of their brace failure.

The reason for failure; joint bolts slipped

before brace buckling. In general design

practice, however, it is not realistic to design the joint bolts to slip prior to brace buckling

.

5. SUMMARY AND FUTURE PLAN

The following results are obtained from the eleven girder tests:

(1) The steel truss girder is usually non- ductile, but rather large deformation

capacity can be anticipated if the slenderness ratio of each individual member is selected small.

(2) Generally the steel truss girder of brace

failure type is non-ductile because it is accompanied with sudden drop in the load carrying capacity. But after the reduc- tion in strength by brace buckling, the load carrying mechanism of the truss changes into frame mechanism by chord

members, and the hysteresis loop becomes

stable. Relatively large deformation

capacity is expected in cases where slippage occurs in the joints.

(3) Steel truss girder of chord failure type should be restrained from its lateral buckling to ensure the deformation

capacity.

(4) Overall steel truss deformation can be estimated as the sum of the deformation of each individual member, of which the truss

is composed.

(5) The inelastic behavior of steel truss girders are largely affected by the

431
plastic deformation of a few individual members near the end of the girder.

In the future, the axial behavior of the member

under various end support conditions should be studied to correctly estimate the deformation

capacity of the truss. Moreover, the evaluation method of the deformation capacity of the frame should be studied through experiments and

analysis. Areas requiring further studies will

be addressed in a future plan.

6. ACKNOWLEDGEMENTS

This research was carried out under the sponsorship of the Building Research Institute

of Japanese Ministry of Construction and the Japan Federation of Architect Office

Association.

The authors are grateful to Dr. H. Yamanouchi, Head of Structure Division of the Building

Research Institute, for his kind review and helpful advice.

The authors also are thankful to Messrs T. Endo and H. Sakai of the Building Research Institute

and to Messrs J. Fukushima, T. Shinagawa, A. Ishikawa, Y. Ito, H. Hashimoto and K. Yamamoto

of the Japan Federation of Architect Office

Association.

(1) Japanese Building Standard Law and its Enforcement Order, The Outline of the Standard is Simply Summarized in "Intro- duction to the Building Standard Law," The Building Center of Japan, 19P6.

(2) Suzuki, H. and H. Nabana, "Behavior of Rigidly Joined Steel Truss Beam Part 2 Snap Through Buckling of a Truss with Weak

Diagonals," Proceedings of AIJ Annual

Meeting, 2373, Oct. 1989 ( in Japanese)

.

432
TABLE 1 DIMENSIONS OF TEST SPECIMENS (SERIES 1)

specimen chord and brace

shear length depth

( cm) angle

(deg) slenderness

ratio#

(cm) chord brace

T-66-21 chord: 2L-65x65x8 Ac=19.5cm

ixc=1.94cm iyc=3.04cm

200 50 66 18 T-66-40 200 90 66 46 T-99-40 300 90

60

99 46 T-132-40 brace:

2L- 50x50x4

Ab=7, 78cm

ixb-1. 53cm

iyb=2 . 38cm

400 90 132 46 T-132-60 400 130 132 76 T-165-60 500 130 165 76

# Slenderness ratio of chord member is calculated as the division of the shear length by the radius of gyration, iyc. The effective buckling length of chord member will be obtained by multiplying 0.73x0.7 to the value in the table.

In the calculation of the slenderness ratio of brace, the buckling length of the brace is considered to be the length between both middle bolts of the end-connections.

TABLE 2 MECHANICAL PROPERTIES

test series member

tensile strength

(tonf/cm^

)

yield

strength

(tonf/cm^

)

L-65x65x8 4.46 2.93

1 L-50x50x4 3.96 2.88 G.PL-9 4.20 2.86 H-194X150 flange

x6x9 web

4.93

5.03

3.45

3.50 H-150xl50 flange

x7xl0 web 5.01

3.90

3.10

2 H-148xl00 flange

x6x9 web

4.88

5.09

3.30

3.50 P-89. 1x4.

2

3.97 3.63 P-89. 1x3. 4.73 4.08

P-60. 5x3.2 4.25 3.92

P: circular tube

433
TABLE 3

DIMENSIONS

OF

TEST

SPECIMENS

(SERIES

2)

OJ

tn n « £ D 0) c E

S-l UJ T3

n!04 <U E- Xi0

cn n T3 o 0) 0O T3 i § o u

434
ENERGY

ABSORPTION

CAPACITY

OF

SERIES

1

TEST

SPECIMEN

435
436
reaction

wall

connected

to

the

/reaction

wall

FIG. 2 TEST SPECIMEN (SERIES 2)

FIG. 4 DEFORMATION MEASUREMENTS (SERIES 2)

437
1000

«

depth

FIG. 3 TEST SET-UP (SERIES 2 )

438
shear force specimen T-66-21

FIG. 5 LOAD vs. DISPLACEMENT RELATION

shear force specimen T-66-^>0

FIG. 6 LOAD vs. DISPLACEMENT RELATION

shear force specimen T-99-W) shear force specimen T-132-40

FIG. 7 LOAD vs. DISPLACEMENT RELATION FIG. 8 LOAD vs. DISPLACEMENT RELATION

shear force specimen T-132-60 shear force specimen T-1 65-60

displacement

8 (cm) FIG. 9 LOAD vs. DISPLACEMENT RELATION FIG. 10 LOAD vs. DISPLACEMENT RELATION

439
shear force Q (ton)

r40

tensitc yiXdina of CO-1 L /-V local bucklinq of CU-1

30 -15 / / . 0 i 15 30 /

displacemenf ./20 ^ specimen A ^ local bucklino of CD-I -AO FIG. 11 LOAD vs. DISPLACEMENT RELATION

440
150 rlOO axial force specimen A

FIG. 16a AXIAL FORCE vs. AXIAL

DEFORMATION RELATIONSHIP

axial force specimen A

FIG. 16c AXIAL FORCE vs. AXIAL

DEFORMATION RELATIONSHIP

axial force N (fon)

50 specimen A B-1

-1 0 1 2 axial displacemenf _25 d(cm)

-50 FIG. 16e AXIAL FORCE vs. AXIAL DEFORMATION RELATIONSHIP

axial force 50

specimen A

N (ton)

f

-2 1faxial displacemenf

1 -25 d(cm)

-50

FIG. 16f AXIAL FORCE vs. AXIAL

DEFORMATION RELATIONSHIP

441
reaction

wall

Mp = 359 t-cm

Mp=449 t-cm

= 4 x 359/150 = 9. 6 ton Q 2 = 4x 359/230= 6. 2ton

= 4x449/116 = 15.5 ton

FIG. 17 OBSERVED HINGE PATTERNS OF BRACE BUCKLED SPECIMENS

442
FIELD INVESTIGATION OF SEISMICALLY DAMAGED SITES BY USE OF

VIBRATORY CONE PENETROMETER

BY Yasuyuki Koga*, Yasushi Sasaki**, Yoshihiro Itoh***, and Osamu Matsuo***

ABSTRACT Liquefaction investigations were carried out at various sites using a newly developed vibratory

cone penetrometer (VCPT) . The results obtained

by these investigations were compared with the cyclic triaxial test results on undisturbed

samples taken from these sites, and the effect

of overburden pressure, etc. were examined.

The liquefaction strength obtained from VCPT

and cyclic triaxial tests at a seismically

damaged river dike were compared.

1. INTRODUCTION

In Japan and the United States, the standard

penetration test (SPT) is most generally used

as an in- situ test method for assessing

liquefaction potential of soils during earthquakes. But there can be a problem that the blow counts (or N-values) from the SPT are influenced by the quality of the bore hole and

it is not necessarily appropriate to investigate the liquefaction potential at many points

using the SPT in structural foundations with

wide area or long length.

To solve this disadvantage, a method using the vibratory cone penetrometer (VCPT) was devel- oped at the Public Works Research Institute

(PWRI) to supplement the SPT in assessing

liquefaction potential at a site. Field and

indoor tests using this apparatus have been

carried out. The VCPT and some tests results were introduced

in the past UJNR Joint Meeting^

.

This paper presents the estimation formula of liquefaction potential of the ground by use of

VCPT. Also, a comparison with detailed

investigation and analysis of the damaged dikes

due to liquefaction in the Nihonkai-chubu

Earthquake is described.

*Head, Soil Dynamics Division, Construction

Equipment and Method Department, Public

Works Research Institute

**Head, Ground Vibration Division, Earthquake

Disaster Prevention Department, do ***Researcher , Soil Dynamics Division, Construction Equipment and Method

Department, do

2. OUTLINE OF VIBRATORY CONE PENETROMETER

The vibratory cone penetrometer was developed

to satisfy the following objectives;

(1) to obtain liquefaction strength of soils along the depth,

(2) to record liquefaction strength of soils accurately including loose silty fine sands with fine contents,

(3) to precisely measure liquefaction strength

of soils at sites,

(4) to obtain liquefaction strength of soils

at sites rapidly, and to easily record

data at many sites,

(5) to perform in-situ tests inexpensively.

The VCPT apparatus is similar to the Dutch Cone Penetrometer. A basic difference, a small electric vibrator is installed inside the vibratory cone as illustrated in Fig. 1. When the vibrator is turned on, the vibratory

cone body vibrates around its axis. By this excitation, the soil's pore water pressure surrounding the cone may build up as in the case of saturated sandy ground. Therefore, penetration resistances under the condition

using the vibrator will become smaller than

those without excitation. Although the effects of the vibrator on soils surrounding the cone might be different from

actual earthquakes, the pore water pressure

increase is an indicator of vulnerability of

the in-situ soils against liquefaction. The difference between penetration resistances with

and without cone excitation will reflect in- situ liquefaction resistance of soils.

Fig. 2 shows an example of in-situ measurement

of penetration resistances. As shown in this

figure, the penetration resistances decrease

when the vibrator is used. This difference can

be expressed by the degree of penetration

resistance decrease D which is defined by the following equation.

Q! - D (1) qs

443
where D degree of penetration resistance decrease

Q| : penetration resistance without operation of vibrator (kgf/cm^ )

Q^: penetration resistance with operation of vibrator (kgf/cm^).

The value of D, the degree of penetration

resistance decrease is nondimens ional and can

have any value between zero and one. When soil

has value of D close to one, the soil can be regarded vulnerable to liquefaction during

earthquakes

The relation between Rq and D was examined

assuming it is influenced by the overburden

pressure as the vibratory load is constant

along the depth as mentioned above. Fig. 4 shows the relation between D and the liquefaction strength (stress) rg with different marks according to overburden pressure. The figure

shows that the relation between D and rg may be separated as groups according to overburden

pressure. The experimental formula which

relates D, rg and a^' can be in several different equation forms. Eq. (2) below is a tentatively proposed formula; Table 1 summarizes the main features of the vibratory cone.

3. LIQUEFACTION STRENGTH AND DEGREE OF PENETRATION RESISTANCE DECREASE

The VCPT measures a penetration resistance

(bearing capacity) in the ground with the application of an external vibration. The dynamic shear stress which is generated in a ground during earthquakes is considered to increase roughly in proportion with the depth

of overburden pressure. Therefore the cyclic

shear stress is controlled roughly to be proportional to the overburden pressure in a cyclic triaxial test which is widely used as a laboratory test method. Meanwhile the magnitude of the acting vibration force in the -VCPT

is considered almost constant irrespective of

the depth. This is the same as in the case of

the SPT where a constant falling energy is given irrespective of the depth. Therefore it

is expected that the penetration resistance

decrease D which was introduced as Eq. (1) to represent the liquefaction potential may be influenced by the overburden pressure under the condition of a constant liquefaction strength Rg (stress ratio). In other words, the penetration resistance decrease D may not necessarily

uniquely correspond to the liquefaction

strength Rg.

Fig. 2 shows a test result at Showa-Ohashi in Niigata prefecture where relatively uniform

loose sand layer is deposited. This figure illustrates that an average value of penetration resistance D decreases as the depth

increases

Table 2 summarizes the vibratory penetration

test results where the cyclic triaxial test results of undisturbed samples are available.

In Fig. 3 the relation between the penetration

resistance decrease D and the liquefaction

strength (stress ratio) Rg is plotted with

different marks according to overburden pressure. The solid curve shows an experimental

equation between D and Rg, which was proposed

during the 16th UJNR meeting. Fig. 3 shows that though the obtained results are distributed

around the curve, the scatter is fairly large.

rg - 0.2a.y'2.(l-D)2 0.144

0.3 (2)

where a^'effective overburden pressure

(kgf/cm^ )

rg ; estimated liquefaction strength

(stress) (kgf/cm^).

Fig. 5 illustrates a comparison between rg calculated from Eq. (2) and the measured

values. The figure shows a fairly good agreement where rg is small (say rg <0.3

kgf/cm^ ) but the scatter is large where rg becomes large.

To examine the cause of scatter between the estimated value from Eq. (2) and the measured

value of liquefaction strength, the difference between them and grain characteristics are compared in Fig. 6. Rg20 * ^ calculated

liquefaction strength (stress ratio) using Eq. (2), and Rg2 o ^ measured value. Fig. (a) shows

that the difference decreases as the mean

diameter Djg increases, though the scatter is large, which suggests that Eq. (2) provides a little smaller value where Djq gets small.

Fig. (b) shows the relation between the difference and the fine grain content FC; a clear relationship is not observed.

4. INVESTIGATION OF A DAMAGED SITE BY USE OF THE VCPT

The Nihonkai-chubu Earthquake of May 26, 1983 caused embankment damages in many places from

liquefaction. Several damages to Iwaki-gawa

river dike in Aomori prefecture were observed

among the various damaged fill sites. Soil

tests and surveys including cyclic triaxial

tests on undistributed samples were performed

using the VCPT.

A detailed case study based on test results

were performed to clarify the damage mechanism.

Fig. 7 shows a soil investigation result of a damaged river dike where settlement of 20-50 cm

occurred. An analysis was conducted to calculate a seismically induced permanent deformation of a dike. A similar analysis method was used as a degradation of a static rigidity.

Following two strain potential
characteristics; (1) liquefaction strength

under the condition without initial shear before earthquakes, (2) cumulative strain characteristics under superimposed stress conditions of initial and seismic shear stresses before and during earthquakes. The latter characteristics were obtained from

cyclic triaxial test results. The former characteristics (liquefaction strength) were estimated with emphasis on cyclic triaxial test results on undisturbed samples from the site. Using both characteristics a model for the analysis was constructed. Liquefaction

strength (stress ratio) Rgao cyclic

triaxial test results, that for analysis, and

that obtained from the VCPT are shown in Fig.

7. This shows good agreement of the liquefaction strength from the VCPT and that of the model for the analysis. The latter indicates

more subtle change of ground strength.

Fig. 8 illustrates a calculated permanent

deformation of a dike against the earthquake.

For the dike crest settlement, the observed

value from a damage investigation is 20-50 cm, the calculated value is 46 cm, which indicates

that the analytical method and properties used

are both rational. This further suggests that

the VCPT can give an in-situ liquefaction

strength with good approximation.

5. CONCLUSION

The following concerns the VCPT which was developed at PWRI;

(1) It is necessary to consider the effect of overburden pressure to estimate the liquefaction strength of the ground from

the measured penetration resistance

decrease

.

(2) An experimental formula to give the liquefaction of the ground from the measured penetration resistance decrease

was tentatively proposed.

(3) The above experimental equation gives a little smaller value when the average diameter of the soil gets small.

(4) The liquefaction strength was estimated

from the VCPT at the site where a dike was seismically damaged. A detailed ground

investigation and analysis was conducted. There were good agreement with the perma- nent deformation analysis and with the measured settlement.

6. REFERENCES

1)

Sasaki, Y. and Koga, Y. "Vibratory Cone Penetrometer to Assess the Liquefaction

Potential of the Ground," 14th Joint

Meeting. UJNR, 1982.

2) Sasaki, Y. , et al., "U.S. -Japan Cooperative Research on In-site Testing Procedure for Assessing Soil Liquefaction (No. 1)," 16th Joint Meeting, UJNR, 1984.

3) Sasaki, Y. , et al., "In-situ Test for Assessing Liquefaction Potential Using

Vibratory Cone Penetrometer," 17th Joint

Meeting, UJNR, 1985.

4) Lee, K. L. , Seismic Permanent Deformations

in Earth Dams, Report No. UCLA- ENG -749 7

,

School of Engineering and Applied Science, University of California at Los Angeles,

1974.

445
Depth

(in)

3) Porous metal

d) Pore pressure transducer

3) Load transducer of point resistance

3) Load transducer of side friction ® Take-out cable for transducer ® Vibrator 7) Power source cable for vibrator ® Penetrating rot

Fig. 1 Vibratory Cone

N-Value Penetration Resistance Qc (kgf/cm’)

Fig. 2 Example of In-situ .Measurement of VCPT Showa-Ohashi Bridge Site

446
Stress

Ratio

of

Cyclic

Iriaxial

lest

to

Cause

Licjuefacti

at

20

Cycles

Ri

20

Degree of Penetration Resistance Decrease i 1-D)

Fig. 3 Relation between Liquefaction Strength and Degree of Penetration Resistance Decrease

447
Stress

of

.'Cyclic

Trlaxial

Test

to

Cause

Liquefactii

at

20

Cycles r

1

o(kgf/cm-)

Degree of Penetration Resistance Decrease (1-D)

Fig. 4 Relation between Liquefaction Strength and Degree of Penetration Resistance Decrease

448
sured

Stress

of

Cyclic

I'naxial

Test

to

Cause

l.iquefaction

at

20

Cycles T

«

20

(kgf/tm’)

Calculated Stress to Cause Liquefaction at 20 Cycles r » 20 ’ (kgf/cni’)

Fig. 5 Comparison of Measured and Calculated Stress to Cause Liquefaction

449
a) Effect of Mean Diameter

Fine Grain Content FC (%)

b) Effect of Fine Grain Content

Fig. 6 Effect of Grain Size Characteristics

450
Resistance

Fig. 7 Ground Investigation Result

, Scale of

I Displacement 5m

! 451
Table 1 Vibratory Cone

a I Comparison of Vibratory Cone

Standard

1 Vibratory Cone

Cone II VI

Diameter mm ) 35.7 41 44

Length mm205 789.6 743.8 Apex Angle 60“ 60“ 60“

Weight kg) 1.1 6,3 6.7

Vibrator Without With With

Sensor Without Without With

b> Specification of Vibrator

Item

Size, etc.

II VI

Diameter 41mm 31mm

Length 576mm 431.5mm

Frequency 200Hz

Power Source AC 30 48V 200Hz

Centrifugal Force 80kgf 80kgf

' c ' Specification of Transducers

Symbol Item Type

Capacity

II VI

qc Load Transducer

of Point Resistance

Strain Gauge Itf 5tf

fc Load Transducer

of Side Friction Strain Gauge Itf —

P

Pore Pressure Transducer

Strain Gauge 2kgf/cm^ 5kgf/'cm“

S

Displacement Transducer of Penetration Potentiometer 1 06cm

452
Table 2 Summary of VCPT & Cyclic Triaxial Test Result

SITE

z(m

'

Qc‘

kgrcnF ) Qc'

kgf C!tf) D kgf L-tfi R I 20 Zi20kgf ./CHI' N(t

D50 mm ; FC

- 1 A- 1- 1 2.28 64.00 12 00 0.813 0.27 0.290 0.076 5 0 o.;:o 4 A- 1- 2 3.33 35.00 30 00 0.143 0.36 0.400 0, 144 . 1 A- 1- 3 4.73 84.00 67 00 0.202 0.49 0.410 0.2C1 3,2:: 12 0 A- 2- 1 3.31 42.00 12 00 0.7H 0.18 0.330 0 125 •*.290 2.0

A- 2- 2 4.85 31.00 5 00 0.839 0.52 0.325 0.169 5 3 0 . 240 0.5

A- 2- ] 10.80 80.00 23 00 0.713 1.05 0.290 0-294 7 -1 JCQ

A- 3- 1 3.77 60.00 94 00 0.000 0.53 0.365 0.193 3 .3 0 33C 4 0 A- 4- 1 4.08 35.50 6 00 0,831 0.48 0.347 0.16? 0 250 1-0

A- 4- 2 2.96 12.80 ? 60 0.797 0.38 0.272 0.103 0.270 2.5

A- 5- 1 6.42 78.00 3 40 0.956 0.75 0.343 0,257 0.310 2.0

A- 6- 1 3.41 43.60 20 0.743 0.50 0.291 0.146 0 :59 3.0 0.230 12 (1 A- 7- 1 5.09 108.80 5? 40 0.518 0.65 0.316 0.205 0 200 5.5

A- 7- 2 4.84 99. 10 36 30 0.634 0.63 0.353 0.22? 0. l^fi 57.0

B- 1- 1 1.80 59.00 ? 00 0.966 0.25 0.304 0.076 3 9 0 350 1.0

B- 1- 2 4.85 44.00 14 00 0.682 0.53 0.368 0.195 5 0 0.45(1 0.5

3- 1- 3 6.55 46.00 14 00 0,696 0.68 0.310 0.211 0 0.400 1.0 B- 1- 4 9.84 26.00 26 00 0.000 0.98 0.315 0.309 3 4 0.26(1 l.O B- 2- 1 2.85 40.00 50 0.988 0.35 0.348 0,122 3 3 0 058 53.0

B- 2- 2 4.43 55.00 0 50 0.991 0.49 0.233 ' 0.114 5 8 13.229 20,5

B- 2- 3 5.'37 65.00 4 00 0.938 0.58 0.215 0,125 9 0 0.250 13,5

B- 2- 4 7.32 75.00 6 00 0.920 0.77 0.217 0.167 12 0 19,0

B- 2- 5 8.13 91,00 13 00 0.857 0.84 0.238 0.200 11 0 0.240 23.0

B- 2- 6 9.38 72.00 16 00 0.770 0.96 0.210 0.202 10 0 0.245 19.0

B- 2- 7 11.38 92.00 36 00 0.609 1.15 0.190 0.219 10 0 0.295 15.0

B- 3- 1 2.43 55.00 4? 00 0.354 0.32 0.528 0.169 3 8 0.038

B- 3- 2 3.18 17.00 10 00 0.412 0.38 0.268 0.102 2 0.072 51,0

B- 3- 3 5.04 68.00 50 00 0.265 0.52 0.275 0.143 23 0 0,225 18.5

C- 1- 1 16.50 40.92 42 74 0.000 1.57 0 . 272 0 427 4 0 0.330 13.0

C- 2- 1 2.38 15.04 3 .00 0.001 0.36 0.410 0,146 3 1) '3.024 83.0

C- 2- 2 8.47 81.30 47 . 04 0.421 0.91 0.330 0.299 14 0 0.100 38.0

C- 3- 1 5.45 121.04 98 37 0.187 0.60 0.435 0.261 15 0 0.240 5.0

C- 3- 2 9.43 101.58 90 74 0.107 0.95 0.385 0,367 12 0 0.14C 21.0

C- 4- 1 3.27 45.64 28 82 0 369 0.35 0.340 0.119 9 0 0.080 49.0

C- 4- 2 9.35 93.38 97 90 0.000 0.90 0.305 0 274 10 0 0.105 32.0

C- 5- 1 6.71 55.30 25 10 0.546 0.96 0.255 0.245 0.32(3 9.8

C- 5- 2 8.25 96.60 34 40 0.644 1. 10 0.352 0.387 0.340 5.0

C- 5- 3 12.30 205.40 123 20 0.400 1.47 , 0.345 0.507 0.350

C- 5- 4 15.38 90.80 64 50 0.290 1.74 0.290 0.505 0.067 52.6

C- 6- 1 4.38 43.00 5 10 0.801 0.59 0.270 0.159 0.340 3.7

C- 6- 2 7,88 120.90 lOl 60 0.160 0.90 0.310 0.27? 0.390 2.8

C- 7- 1 2.35 71.50 45 50 0.364 0,46 0.260 0.120 0,430 2.3

C- 7- 2 3.92 41.30 52 20 0.000 0.55 0.290 0.160 0.570 1.9

C- 7- 4 8.33 165.30 81 40 0 508 0.94 0.235 0.221 0 280 9.3

C- 7- 5 10.62 167, 10 160 20 0.041 1.15 0.280 0.322 0.130 23.4

C- 8- 1 8.63 206.00 91 00 0.558 1.02 0.260 0.265 12 0 0.290 9.0

C- 9- 1 10.95 114.00 123 00 0 000 1.11 0.322 9 0 0.100 38.0 C-10- 1 6.75 149.00 150 00 0.000 1. 16 0.220 0.255 9 0 0.370 4.0

0- 1- 1 5.73 41.20 32 50 0.211 0.63 0.230 0. 176 6 0,033 74.3

0- 1- 2 8.18 67.00 74 80 0.000 0 32 0.290 0.218 13 fl 0.260 5.4

0- 1- 3 9 33 113.70 102 00 0.103 0.92 0.300 0.276 14 0 0 440 5.2

B- 2- 1 2.46 17.10 11 40 0.333 r'.4l 0 315 0.129 1 0 0.096 41.2

D- 2- 2 3.51 13.50 7 00 0.481 0,305 0, 146 2 0 0,136 25,0

0- 2- 3 5.56 71.70 24 90 0.653 0. 6.1 0.270 n.PO 10 .0 0 490 3.4

D- 2- 4 7.21 101.90 80 10 0.2L4 0.73 0,290 0,226 17. 0 0.265 5.0

E- i- 1 4.8d 54. 10 51 oo 0 137 os 0,200 0.160 0.440

F- 1- 2 5.83 70.80 52 so 0.254 0.57 0.368 0.210

F- 1- 3 6.53 93.30 75 90 0.186 0.63 0.278 0.175

F- 2- 1 6.94 52.40 15 0.330 0 65 0 173 0 242

F- 3- 1 6.94 133. 5U 122 20 (J.0S5 0-6? 0 373 0. ;42

F- 4- 1 5.44 48.50 43 30 0 107 0.52 0.400 0.208

F- 4- 2 7.34 72.60 60 90 0,161 0 61 0 383 0.264

F- 5- 1 8.34 58.30 34.20 0.413 0 78 0.345 '),269 F- 5- 2 10.30 146.80 128 90 0.122 0.575 0.552

F- 5- 3 12.34 122.30 96 10 0.214 1.14 0.407 0 464

F- 6- 1 3.33 13. 'O 21 00 0 377 0.33 0.27'' 0. 124

F- 6- 2 12.53 123.90 104 no 0.161 1. 18 0 483 0.560

F- 7- 1 3.45 36,70 53 70 o.ooc 0.35

F- 7- 2 7.31 IH 50 102 10 V :08 0 69 0 350 0 242

F- 7- 3 9.34 150. 10 143 80 0 '142 (1 38 0.510

F- 8- I 5.25 42 10 38 40 0.088 ,. 0 450 0 225

F- 3- 2 6. 10 72,20 72 10 0.001 0 58 0 442 0.256

C- !• 1 14 43 24 60 1? 40 n.:03 0 83 D 307 0.255 3.6 0.042 60.0

C- 1- 2 2? 30 ,00 c 560 ') 110 0 353 3 0 072

C- 1- 3 19.35 40 30 .10 135 1

' 295 0.360 ». 8 0.115 36.0

453
AN INVESTIGATION ON A TWELVE-STORY OFFICE BUILDING DAMAGED BY THE 1985 MEXICO EARTHQUAKE

by Hisahiro Hiraishi

Masaya Hirosawa

Hldeo Tsukagoshi

Toshihide Kashima

Toshikazu

ABSTRACT

The 1985 Mexico City Earthquake caused serious

damage particularly to reinforced concrete

buildings. A team of experts was dispatched to Mexico October 19 to November 22, under sponsorship by the Japan International Cooperation

Agency. Their objective was to evaluate damaged

buildings to judge the necessity for repairing

and strengthening and to propose a plan for strengthening appropriate buildings.

This report describes the detailed damage obser- vation and analyses and a plan to strengthen a twelve story reinforced concrete office

building.

The following conclusions were derived from the observation and analyses:

1) Torsional vibration occurred during the earthquake due to the existence of exterior

brick walls.

2) Significant settlement occurred and shear walls did not perform as expected because of insufficient bearing capacity of friction

piles.

3) A wrong location of the cut-off of top reinforcements of beams resulted in a critical section occuring at a different

location in the structural design. As a result, the beams' flexural strength was extremely small.

4) The base shear coefficients corresponding

to the ultimate stage were 0.07 and 0.08 for longitudinal and transverse directions, respectively (due to 2 and 3 above).

5) Consequently, a drastic increase in strength

was needed to strengthen the building.

1.

INTRODUCTION

The 1985 Mexico Earthquake with a magnitude of 8.1 caused serious damages to Mexico City

buildings. Although damage statistics are not

yet available, it is estimated that 7,000-

8,000 lives were lost, 400-500 buildings totally

collapsed, and about 1,500 buildings were evaluated for necessary repair and strengthening .

Kawashima-*- The objective of the team of experts dispatched

to Mexico City during October 19 to November 22 was: to evaluate buildings damaged, to decide

the necessity for repair and strengthening, and

to propose a plan to strengthen buildings.

This paper describes the investigation and analyses of a damaged twelve-story office

building.

2. OUTLINE OF THE BUILDING

The 12 story building had a trapezoid plan with

5 spans in the longitudinal direction and 3 spans in the transverse direction and a basement,

as shown in Fig. 1 and Photos 1 and 2. The basement and the first and second stories were

used for parking. The structural feature of the building is as follows

:

1) The building had the center core consisting

of one and two shear walls in the longitudinal

and transverse directions, respectively.

2) The columns were flat section in the longi- tudinal direction, and the flat ratio was 1/2.

3) Floor system was Losas Planus (Mexican style waffle slab); beams and sub-beams were placed within the slab.

4) Brick walls reinforced by reinforced concrete

braces were wholly placed in two exterior

frames out of four exterior frames.

5) There were one in-situ reinforced concrete friction pile and two piles, under interior

and exterior columns, respectively, having a triangular section of 55 cm x 55 cm x 55 cm, and 26m long.

6) The Building was constructed during 1980 to 1982.

3. DAMAGE OBSERVATIONS

3 .1 General Description of Damage

^Ail authors from Ministry of Construction,

Building Research Institute, Tsukuba, Japan

454
The general description of damage is summarized

as follows:

1) Failure of beams connected to non-structural

reinforced concrete walls on the north side

of the building and the adjacent slabs.

2) Cracking on the floor slabs due to settlement of the elevator core.

3) Failure of capitals on the exterior columns

on the east, north, and south sides.

4) Failure of short span beam between the C and

D frames on the south side.

5) Failure of extremely short beams in the basement.

6) Failure of columns and beams at the anchorage of reinforcements in the exterior brick

walls (east and south sides).

7) Failure of exterior brick walls.

8) Failure of lumped slabs in the parking space. Torsional vibration is considered to have occurred during the earthquake due to the existence of exterior brick walls on the east and south sides and the lumped slabs for parking in

the first through third floors. Bond splitting

cracks were observed in columns.

3

.2

Structural Damage

Significant beam damage was seen at the capitals

except for damage of the beams between the frames

1 and 2 which were connected to the non- structural reinforced concrete walls on the north

side and for damage of the short span beams in

the frame 1 between the frames C and D. Large

column damages were observed at the anchorage of reinforcements in the exterior brick walls except

in the basement. Some damage examples on columns

and beams are shown in Photos 3-5. Here, notation S, F, B mean shear, flexure, and bond splitting failures, respectively, and notation

II, III, IV, and V mean small, moderate, severe damages, and collapse, respectively. Observed

crack patterns on the columns and shear walls on

the frames 1, 3, A, D, and F are illustrated in

Figs. 2-6. Cracks running at 45 degrees in both directions on shear walls indicated that they carried seismic forces reversibly. Examples of crack patterns on the 10th floor are given in Figs. 7 and 8. The following statements

are from the crack observations.

1)

On floors above the first floor, cracks of 3.0mm in width run parallel to the frames B and E at the locations 1 ,500mm-l ,800mm apart

from the frame lines. These locations correspond to the end of capitals where the amount of upper ref inf orcements suddenly decreased and reinforcing bars yielded. These

cracks resulted from settlement of the elevator

core

.

2) Cracks of 3.0mm in width running perpendicular to frame A also were seen between

frames 1 and 2. 3) On the lower side of floors, radiating cracks

from the column top were observed.

3 .3 Non-structural Damage Damage on brick walls in the frame lines was investigated. They were classified into 6 cate- gories as shown in Fig. 9. Typical patterns of failure for each type of brick walls are shown

in Table 1. Exterior walls of east and south facades were

type X, and most were damaged as shown in Photo 4. The detail of anchorage of reinforced

concrete braces of x type is shown in Fig. 10; such anchorage caused severe damage to beams (damage to columns if anchored into column), as seen in Fig. 4. The structural engineer who designed this building said that the x type brick

wails were considered an earthquake resisting member in structural calculation. However, the efficiency of the brick walls is quite doubtful

for such buildings in which ductility is required, although it may be effective for stiff buildings. Crack patterns on the lumped slabs in the parking space are illustrated in Fig. 11. As large cracks were observed specially between the frames D-F, the lumped slab between the frames

1-4 and frames E-F were considered to have acted

as a brace, and the slab between the frames 1-2 and E-F were considered to have acted as a flexural member. This fact may explain that the building vibrated in torsional modes to produce

severe damage to the fourth through sixth stories rather than to the first through third stories.

The non-structural RC-walls on the north side were located outside of frame A. The columns at each end of the walls were connected to the corner column of the frame 1 or the beams on the frame A between the frames 1 and 2 (see Fig. 1). The foot of the wall was connected to the retaining wall at the basement just outside of frame A. The shear cracks on the walls indicate possibly that large seismic forces were carried

by this wall. Concrete crushed at the foot of

the wall and shear failure in the beams on the frame A between the frames 1 and 2 and the large cracks in the floor systems around the wall were

induced by the failure of the wall (see Photo 6).

3 .4 Collision

This building collided with the adjacent buildings in the east and south sides (see Photo 7). The building on the south side was particularly

damaged.

455
3.5 Settlement and Tilting of the Building

Settlements of the floor systems were measured at the basement, first, tenth, and twelfth floors by bench mark leveling. The largest settlements occurred around the elevator core. The inclinations of the building were measured by transit to determine the considerable distortion of the building (see Fig. 13).

4. STRENGTH TESTS OF CONCRETE

Strength of concrete was measured by the Schmidt

hammer tests at 23 points in each floor and the tests on the three concrete core samplings taken

from the column at the same location for all the stories. The average strength of concrete estimated from the Schmidt hammer tests was

205 kg/ cm2 for the whole building. The same quantity measured by sampling tests varied between 210 and 400 kg/cm2. The average strength of concrete at the middle part of columns was 248 kg/ cm2

.

5. ESTIMATION OF FAILURE MECHANISM

Fig. 14 shows the estimated failure mechanism

from the crack patterns, where the cracks of the width of 1mm and the length of 50cm were con- sidered to be the threshold value for the hinge occurence. Shear failure was not observed in this frame. Notation o and x mean flexural yielding hinges corresponding to each directional response, respectively. For the frames

1 and 3 , the hinges x appear at many locations between the frames D and F. This is possibly a result from the settlements of the frames C and

D. There are seen, however, the hinges o between the frames A and C and the hinges x between the frames D and F, suggesting possible

severe oscillations of the building in both directions. Since the hinges appear in more

than half of beams in the transverse direction,

the building is considered to have oscillated severely in this direction.

6. EVALUATION OF LATERAL STRENGTH AND SEISMIC

CAPACITIES

Bearing capacities and seismic safety estimated

by the third Level Screening Procedure for eval- uation of seismic capacity are described for the undamaged state of the building. In calculating

bearing capacity of the building, two methods

were employed. In the first method, the nodal moment distribution method for limit analysis

was applied to the building whose base was fixed

at the basement. In the second method the virtual displaement principle was used for evaluating bearing capacity of the building with

its base fixed at the ground level. The horizontal strength of frame structures and shear

walls were first calculated separately and then

added to give the total bearing capacity of the building

.

Lists of structural indices (Is=Eo x SD x T) are

tabulated in Table 3. The values for the first were 0.061 in the longitudinal direction and 0.082 in the transverse direction.

The bearing capacity was re-evaluated by applying the virtual displacement principle under the assumption of reversed triangle distribution of external forces. The derived story shear forces

are tabulated in Table 4. The base shear coefficients corresponding to the ultimate state,

are 0.07 and 0.08 for the longitudinal and trans- verse directions, respectively. The low values

of bearing capacities, resulted from cutting of the reinforcements in the upper side of slab- systems at the end of capitals. This action

resulted in extremely low beam strength and insufficient bearing capacity of the foundation

was not allowed to increase the strength of the shear walls. The calculated ultimate lateral

strength, where all reinforcement in the upper

side of the capitals is assumed to be effective,

is twice that calculated based on real condition

as mentioned before.

7.

PROPOSED PROCEDURE OF STRENGTHENING

Figure 18 proposes an example for possible repair

and strengthening procedures. The target level

of strengthening is set at 0.25 of the strength

index and 2.5 of the ductility index being con- siderably higher than the existing bearing

capacity of 0.07-0.08 in the undamaged building. Four shear wails are installed at the corners in both directions, each is expected to carry the base shear coefficient of 0.05. This arrangement of shear walls is effective to prevent

torsional motions of the building. Existing

shear walls are thickened to prevent shear

failure in future earthquakes.

The following local reinforcements of elements

also are recommended:

(1) Non-structural RC walls on the north side should be removed. The surface concrete of the floor slabs near the wall is chipped and recast after the welded mesh wire is placed.

(2) Shear walls are installed in place of the short columns in the basement. Short span beams on the frame 1 between the frames C and D should be improved by removing the brick wall or by providing a slit between

the beams and the brick wail.

(3) Since the target strength of the building

is relatively low, the members and structures should be ductile enough to avoid failure in shear and bond splitting.

456
Fig. 1 Typical Plan

457

U95Q
458
I : Cover concrete htis dropped out

: Crack observation is impossible

R

Fig. 2 Crack Pattern of 1 Frame (West Side)

459
B : Cover concrete has dropped out O : Crack observation is impossible

Fig. 3 Crack Pattern of 3 Frame (West Side)
; Cover concrete has dropped out

El : Crack observation is impossible R

12

' \ / X

11 ><i\ /

/\

10 .>1 \ /

9 \ / X

8 i \ ./ X /

7 X)\ /

[ 1

6 XIXIX

5 xs 1

4 XX

3 xi;XXI

2 XXfX^^'i

1

B1 [ J

4 3 2 I

Fig. 4 Crack Pattern of A Frame (South Side)

461
: Cover concrete has dropped out

: Crack observation is impossible

R

12 Xf

11 >< :XI

10 -XI

9 /\V V :D<[

8 X 1’ X 7 XV

6 X XI

5 X XI

4 X:\ XI

3 X: ^xi

2 X1 >< \ \ l^raanT

B1 X1 1

1 i 4 3 2 1

Fig. 5 Crack Pattern of D Frame

(South Side)

U : Cover concrete has dropped out

: Crack observation is impossible

12 Ix^ X. X X

u IX' x^

10 IX' X\ X 9 X-><

8 XX

;

X' X 7 IX X

X. X:

s iX XI

s IX X

IX X

X X:;

3 xX ><]

2 X X2s : X ^ Xl

X

/ \

^

1

i

1 12i i3 4

Fig. 6 Crack Pattern of F Frame

(North Side)

462
N A B C D E F

Fig. 7 Crack Pattern of 10th Floor System (Upper)

Fig. 8 Crack Pattern of 10th Floor System (Bottom)

463
Table 1 Typical Failure Patterns of

X Type P Type M Type

S Type HI Type H2 Type

Fig. 9 Construction Types of Brick Wall

i ^^Snchore bar 4»2.5

Length of connection yN Reinforcing bar 4#2.5

» 2@30

Brick Wall

Type Daoage Cons tmct ion Type

X

1) Crush at comer

2) Crush at crossing point

3) Collapse of anchored bars

4) Col laps e of connected bar at comer

1)

2)

PM

1) Diagonal crack 1)

S

1) Diagonal crack

2) Cmsh at bottoe

1) ^

HIH2

sane as above same as above

Fig. 10 Anchoring of Diagonal Reinforcing Bars

2F-1F Baiip A B C D E F

Fig. 11 Crack Pattern of Ramp

I)

]

1

]

i)

I]

]

]

:

]

]

]

]

]

464
Photo 6 Failure Pattern of Non-Structural Photo 7 Collision at East Side

R/C Wall at North Side

A C E F

Fig. 12 Settlement of Basement Floor System (Unit:mm)

465
Fig. 13 Permanent Deformation of Floors

lx 0 e .1 0 A

X O

&

lx t 1 0

o X 0

lx 0 lx ir 1 0 X

1 ^

lx ae 1 o

o Q.

lx *i: X j A 0

ro >

li “

r> u :i O

X o

X X o

lx K

ro > r p

o

O L _ A lx ar j

0

o A

1

1

' 110 1 lx II aI X <? KO X

lx ir 1 X

- X

1 L ir 1 X —

X

1 II 1

1 1 II 1

A i i iE F

Fig. 14 Failure Mechanism
Table 2 Seismic Index of Structure

Before Earthquake

>1 COITir

Longi tudinal Span C

T29DD F Is(Eo) C F Is(Eo)IS 0303*1

(0.384) 105.27 LSI 0.402 *1 37U8 1.0 5155 IS 0.362)

IS 0.167179.96 15 — — — — U202

“T3?.8i li 0.107

184.54 1.5 — — — 0.141

9

171.44 1.5 1197:891 1.5 — — 0.102 — — 0.117 8

18021 1.5 0.089 '21958 1.5 — — — — 0.104

7 “18902 1.5 0.082

248.11 1.5 — — — — 0.107

67

19684 1.5 0.077

283.49 1.5 — — — 0.110 ^233 1.5 0.073 29229 li — 0.105 ~mn 1.5 0.071 29968 1.5 — — — — 0.102

3 22u54 1.5 0.072 29011 1.5 — — - — 0.096

2

1.5 0.07 5 298.58 1.5 — — — 0.094

1 196/i4 1.5 0.061

26A.'3a 1.5 — — — — 0.082

Values in parentheses mean values at stage

of beam yielding

Table 3 Lateral Strength

j a)

1 ® (1 Q Cl Q Ct Ct * Ct

IZ ISIJZ 1910I 14T 19193 1.. 9U4 11 4 9199

11

j Tiroo 149.4 9 0.20 1T19M lUO li 3 1131 1..

,0 IniMi 1 3 9.1

1

11 2 19U4 11 9 13116 11 3 .23. 0 1 4

. |M.n. IT 144 at 1 ..... 11 3 1TU4 11 1 70199 11 3

1 ||«TSS4 1 903) 109 210JI 11 1 30133 11 1 ..132 113

I jjsrm 119.02 10. 2411 110 23U4 1.0 290J3 112

1 larii.o 19114 2114 110 29176 109 31139 li 1

j jmios 201.3 lot 2.13 109 2917 10. 13103 11 1 3J9U6 2919 10. 20.,.. 10. 354.11 010

3 399144 10 9 2.1. 107 3110Q 39IJC lot

z 44IU9 zsTJa 10 s 29U1 10 7 12U1 10. 379J! 109

1 4azu< 19144 10 4 2IU9 10 9 32196 394JI 109

(D Nodal moMnt

<S Limit analysis ® Story ® We

stributioi by assumed h ght ® Lon

for limit analysis

ge mechanism

tudinal ® Span

nmnrTT Thickness- increased shear wall for strengthening

Fig. 15 Proporsal of Strengthening

467
PRACTICAL APPROACH FOR DETERMINING STAGE FREQUENCY

BY Lee Bucler and Thomas A. Hardy^

ABSTRACT

The Corps of Engineers has to provide reliable estimates of coastal flooding from surges in order to make sound engineering decisions

regarding the design, operation, and maintenance of various coastal projects. Potential of flooding due to tropical storms (hurricanes) is greater for coastal regions

along the Atlantic and Gulf of Mexico Coasts

of the United States while extratropical

storms (northeasters) affect coastal areas from the mid-Atlantic to Canada. Storm surges

provide a higher water level on which short period surface waves can propagate, subjecting

beaches and structures to wave forces not ordinarily experienced. There is a need for procedures to analyze the impacts of storm

surge and related wave forces for the purpose

of establishing design criteria for coastal structures, providing data for an economic

analysis of coastal projects, or developing evacuation plans.

Use of time-dependent numerical models for simulating water levels and flows in open

coastal waters, estuaries, bays, or lakes has become a standard practice. When these numerical schemes are used conjunctively with

statistical techniques, a modeling procedure

can be developed to estimate the magnitude and frequency of occurrence of storm-induced water

levels in coastal regions. It is not the purpose of this paper to discuss any particular hydrodynamic, wave, or probability

model development, but rather to discuss the practical application of these techniques to coastal regions.

Two flood frequency studies will be used as examples to Illustrate appropriate solution to modeling difficulties. The first is a dualgrid model study to investigate flood levels

on the south coast of Long Island, New York, including open coast, inland bays and flood

prone land areas. Analysis of improvements to provide adequate protection for the area

against storm tides also was conducted. A

second study involves the conjunctive use of hydrodynamic, wave, physical, and probability

models, applied on a single nearshore grid, to provide estimates of flood frequencies and wave overtopping volumes in an area north of Boston, Massachusetts.

1.0

INTRODUCTION

Efficiency in design of coastal protection is becoming more and more important. Development

of coastal regions, costs of damages from

storm-induced water levels, and costs of protection from these waters are ail increasing. Adequate protection for coastal

regions is desired; however, due to monetary

constraints, the amount of water level protection that can be considered adequate

becomes a question for which there is no easy answer. Therefore, inherent in any coastal protection project (new or rehabilitation

effort), there is a need to develop the best possible estimate of the stage-frequency

relationship for the project area, as well as an estimate of the error in this relationship.

When applying numerical and statistical models

to estimate storm-induced flood levels for open coast and inland situations, various

problems are encountered. These include, but are not limited to, selection of an appropriate model basin; procedures for model calibration and verification; appropriate

treatment of wind-wave generation and wave

setup in the surf zone; barrier island overtopping and routing of flood waters into inland bays; wave overtopping of structures; applicable probability models; and optimization of the number of computer runs necessary for meaningful statistical results. This paper focuses on the type of models used

for stage frequency analysis, calibration and verification procedures, and statistical

techniques for reducing computer effort. Two applications are briefly discussed to point

out how modeling techniques and procedures

were applied to efficiently derive stage frequency relationships for coastal flooding.

2.0 MODELS

2 . 1 Storm/Tide Models

The storm surge equations (5, 6) are descriptive of a variety of other long-wave applications in open coastal waters. These equations are familiar with most readers and will not be described in this paper. Many

techniques and grid constructions are available for use in coastal areas. The applications to be discussed later used the Waterways Experiment Station Implicit Flooding

Model (WIFM). The numerical and hydrodynamic

features of WIFM are discussed in Butler (1). WIFM solves the vertically integrated, time- dependent, shallow water wave equations of

^ U.S. Army Engineer Waterways Experiment

Station, Coastal Engineering Research

Center, Vicksburg, MS 39180-0631

468
fluid motion using an alternating direction

implicit finite difference algorithm. The model allows sugrid barriers and utilizes a stretched numerical grid which permits concentration of grid resolution in areas of interest.

In order to model storm surge it is usually necessary to extend the computational grid

past the edge of the continental shelf and into deep water. Since it also is desirable

to have small cell sizes in areas of interest,

a very large number of grid cells may be necessary to model a study area using a single

grid. Consequently, in locations with a wide continental shelf, a two grid system is usually developed. Fine resolution is required in nearshore regions. Development of

a single grid to resolve fine scale nearshore

features and, at the same time, cover the required expanse of the adjacent continental

shelf, would result in a very inefficient

model to simulate hundreds of storms. The outer (global) grid contains an approximate

description of the inland system thereby eliminating possible contamination of nearshore boundary conditions. Boundary

conditions are saved from global grid simulations to drive the inner (nearshore) grid which contains sufficiently fine resolution to describe flood propagation into

shore, through various inlets, into back bays,

and over flood-prone inland areas. This same technology is used when only separate models

for open coast and inland surge are available.

Grid construction will be discussed later in sections on model application.

To insure the appropriateness of all model

assumptions, tests should be made with varying

grid limits, time steps, and sensitive channel depth representation. Sensitivity to model

boundary conditions also should be tested. Having accepted an appropriate grid, the model

should be calibrated and verified for the astronomical tide. Tidal data are usually

more available for checking model performance

or can be collected without much difficulty.

Following model tests for tidal representation, several historical storms should be simulated to compare model water level predictions with observed levels and observed

extent of flooding.

2 . 2 Wave Model s Wind-generated surface waves force an increase

in the time-averaged water level in the surf

zone. The magnitude of the effect for hurricanes can be an increase of 2 meters or more. Thus, in areas where the principal

storm surge is due to wind stress and barometric effects, there is an additional

rise in the water level due to the presence of waves. This wave setup is not essentially

periodic on the time-scale of the surface

waves but varies with the general sea state

and is restricted to the surf zone. The cause

of the rise is the divergence of the wave radiation stress caused by dissipation of wave energy in the surf zone. A number of complications occur in the estimation of wave setup during storms. Primary among these is the estimation of the radiation stress tensor for storm wave

spectra. Since the sea state during a storm

is not well represented by a simple monochromatic wave train, estimation of the tensor involves adequate estimation of the directional wave spectrum in shallow water. This must take into account spectral wave transformations in shallow water including

shoaling, refraction, diffraction, wave

growth, and breaking. The results are highly

site dependent. Both monochromatic and spectral models have been used to model wave

setup in storm conditions.

Wave setup may not be significant in all cases; some problems may involve estimation of flooding due to waves overtopping a structure.

A steady state, shallow water, directional

spectral wave model is used to perform wave simulations in an example of a single grid

application. The required simulations

actually called for the use of a time dependent model but the cost of using such a model was prohibitive. In lieu of a truly transient simulation, the model was run once

for each hour of each event and the resulting

wave climate was taken to be representative of the conditions existing for the entire hour.

2.3 Probability Methods

There are several possible approaches to establishing frequency curves where the scarcity of data in the immediate study area

requires a modeling approach. The two most common are called the historical method and

the joint probability method (JPM). In the historical method, a series of historical

events is recreated with the pertinent data being saved in the necessary locations. In effect, it is like operating a time machine

with the hindsight to know what data to collect and where to collect it. Probability

is assigned to each event by a standard

ranking method. For the JPM, the storm type

is parameterized. For example, hurricane windfields can be defined by three parameters: central pressure deficit, radius to maximum

winds, and forward speed. Then an ensemble of synthetic events is simulated representing

those events which are possible in the study

area. Probability is assigned to individual

events by assigning probabilities to parameter

values which determine that event. If the parameters are independent, then the probability of the event would be the product

of the probabilities of the component

parameters. In the two examples discussed

below, the JPM was used along with a

469
modification 3.0 of the historical approach.

DUAL GRID APPLICATION

A study to investigate the frequency of storm

plus tide flood levels along the coast and within the bays of southern Long Island, New York, was recently completed at WES (3). The study involved development of two numerical models, collection of field data for model calibration, computation of coastal and back-bay stage frequencies, and analysis of alternative barrier and inlet configurations.

A coarse global grid was constructed to cover

the New York Bight from a point south of Atlantic City, New Jersey, to beyond Cape Cod, Massachusetts (and includes New York harbor

and Long Island Sound). The purpose of the global grid was to model large-scale tidal and meteorological events, providing the boundary

conditions for a smaller, high-resolution grid

of the main study area. The high-resolution

nearshore grid covers the south-central shore

and back-bay areas of southern Long Island and

has variable grid size resolution of 270 to 1200 m. Both numerical grids were tested and calibrated for the M2 astronomical tidal constituent, and verified for a mixed tide condition. Bathymetric and topographic information describing barrier island and inlet configurations for five severe historical storms was developed and the storms were hindcasted. Due to the severity of the historical and hypothetical storms to be simulated for developing a statistical data

base, a scheme to model breaching of the barrier island system was implemented.

The JPM approach was used to develop stage frequencies for hurricanes, northeaster, and

tide events along the south shore of Long

Island. Five hurricane parameters were used: forward speed, radius of maximum wind, central pressure deficit, track direction, and landfall point. Probability distributions for each parameter were obtained by a thorough

review of the historical occurrence of hurricanes in the New York Bight. It was determined that three forward speeds, three

radii, five landfall points with three track directions, two track directions for bypassing

storms, and six pressure deficits were needed

to represent the range of hurricane parameters

which could occur. Thus, 918 hurricane

events were modeled, each with its own probability of occurrence (namely, the product

of the probabilities of the component

parameters). The actual number of storm events modeled was reduced. Sensitivity tests determined that water levels resulting from different pressure deficits could be linearly

interpolated. Therefore, modeling hurricanes

with only two values of central pressure

deficits was necessary.

Northeasters were modeled by choosing 27 historical storms which were found to be representative of the 41-year period, 1940 through 1980. Historical data were used to develop a partial duration stage frequency

curve of significant northeaster surge levels

at Sandy Hook, New Jersey. Sandy Hook is at the entrance to New York Harbor. Proba- bilities of simulated storms were assigned

according to how much of the partial duration probability distribution each storm

represented. Global grid simulations were carried out for surge without tide. Tides were incorporated

through the convolution of tidal values with

output of the surge-only numerical

simulation. This procedure is valid in the open Bight area where nonlinear effects of combining surge and tide are negligible. The convolution procedure produced more than 600.000 possible hurricane surge-tide combinations and over 18,000 northeaster

surge-tide combinations. These storm sets were ranked by height, and tnrough the use of their probabilities, stage-frequency

relationships were developed for open-coast

locations throughout the study area. For nearshore grid simulations, it was necessary to reproduce the statistics of the thousands of possible global surge-tide events with a much smaller ensemble of events for nearshore simulation. A random sampling of roughly 6000 hurricane plus tide events was further reduced to three sets of 17 storm-tide

events, with each set chosen to represent a range of storms from moderate to very

severe. Weaker events were not emphasized due

to the dominance of northeasters for return

periods below 25 to 50 years. Tests showed

that 40 northeaster plus tide simulations

could reliably represent the full set of extratropical events. Probabilities of each nearshore event were assigned according to how much of the nearest open-coast stage frequency

curve each event represented. Stage frequencies were developed from the results of these simulations. This procedure replaced

Che typical approach of routing ail Che opencoast surge events into the back-bay areas.

It also provided a clean statistical method to account for the nonlinear effects involved in routing flood waters through inlets and over

the barrier islands. The method also provides

for measuring confidence in Che results (4). The technique used is known as the bootstrap,

and provides an estimate of Che error of Che stage frequency by determining the variation

possible in each component's probability

distribution. This variablity is due to the construction of the distribution from a finite

set of historical events.

4.0 SINGLE GRID APPLICATION

The use of a two grid approach is not always

470
necessary for investigating flood frequency

via numerical modeling. A good example of a single nearshore grid application is a recent

study conducted at WES (2). The study

location is a residential area, just north of Boston, Massachusetts. Over fifty years of tidal stage data (Boston harbor) and wind data

(Boston's Logan Airport) were available. This information was the key to reducing the computational effort in establishing flood frequencies in Broad Sound. A second factor

in adopting the single grid approach was the lack of hurricane influence on the stage frequency curve at Boston.

The purpose of this study was to determine

flood levels in the back-bay reaches of the Saugus-Pines River system as well as develop

the frequency of water volumes caused by storm

waves overtopping seawalls protecting a residential area called Roughans Point. Flooding caused by wave overtopping is an annual occurrence. In order to determine the economic viability of possible protective

measures, it was necessary to establish the frequency of the flood levels caused by wave overtopping for both the existing conditions

and the proposed protective measures.

Wave overtopping is a complex phenomenon which

is governed by a number of sometimes interdependent variables such as tide, storm

surge, wave height and period, and structure

height and slope. To study this problem five models were used in conjunction. A numerical

storm surge model (WIFM) calculated the stillwater level. A numerical spectral wave model estimated the characteristics of the waves which would attack the structures at the water

levels predicted by the surge model. A physical model produced an analytical

relationship for determining overtopping

rates given water level, wave height and period, and structure characteristics. A

flood routing model calculated the maximum

stage in the interior of Roughans Point caused

by each simulated event. A probability model

chose events to simulate, assigned

probabilities to those events and constructed

the stage frequency curves.

If all of the variables which determine wave overtopping were independent, then a joint

probability approach would work well for determining probabilities associated with

specific overtopping volumes. The probability

of a volume would then be the product of the separate variable probabilities.

Unfortunately, storm surge, wind and wave variables are not independent. Therefore, an alternative approach was developed which would

link the probability of an overtopping volume

to historic storm meteorology.

Defining maximum storm surge as the maximum

difference between measured and predicted

water levels during a storm, a 20-year partial

duration series (PDS) of maximum surge was established at the Boston tide gage adjacent

to the study area. A synthetic surge plus

tide event was created by combining a time history of a surge from the PDS with a randomly chosen astronomical tide. This was

done with the assumption that the historic storm which caused the surge was independent

of tide, i.e. the storm could have occurred at any phase of any tide during storm season. A

large array of synthetic surge plus tide events was created by repeating the above process 500 times for each surge time history.

A probability of exceedence curve for the still-water level can be reliably established

at the nearby Boston tide gage due to its long record. A semi-random selection procedure was developed using this curve as a basis for choosing events to simulate and for assigning

probabilities to these synthetic events. In a sense the exceedence curve for the Boston

Harbor gage was used to replace the need for a global grid. In order to judge the accuracy

of the selection process, the total number of events simulated was divided into three

sets. The sets were individually processed, creating three overtopping exceedence

curves. The range of variation among these

three curves gives an estimate of the accuracy

of the process.

For areas where stage frequency curves were computed for the still-water level resulting

from the combination of surge and tide, only

the surge and probability models were necessary. These areas included both open

coast and estuarine locations. The model

showed the inlet to the Saugus-Pines River system efficiently conveyed flood levels to the river flood plains. This meant that the entire study area tended to flood to approximately the same level. For Roughans Point, multiple combinations of possible seawall-revetment structures were modeled. Major differences amoung Che combinations were evident at the lower return periods with combinations of a wide berm

revetment and a cap on Che existing east seawall providing the greatest protection. Ac higher return periods the protection differential offered by Che various structure combinations tended to diminish.

5.0 CONCLUSIONS

Two example flood frequency studies were used

to describe appropriate solutions to various modeling difficulties. Both dual- and singlegrid approaches were discussed. Practical

problems of routing flood waters over protective dune systems were solved. An efficient procedure for selecting a minimal

number of storm plus tides events Co simulate

in a nearshore region was developed.

Innovative use of five modeling systems to

471
compute wave overtopping frequencies was described

.

6.0 ACKNOWLEDGEMENTS

Portions of the research described in this paper were conducted by Mr. Mark D. Prater, formerly of the Coastal Engineering Research

Center and currently at the University of Washington. The research presented in this paper, unless otherwise noted, was sponsored

by United States Army Corps of Engineer

District, New York, and the New England

Division. Permission was granted by the sponsors and the Chief of Engineers to publish this information.

7.0 REFERENCES

(1) Butler, H. Lee, "Evolution of a Numerical

Model for Simulating Long-Period Wave Behavior

in Ocean-Estuarine Systems," Estuarine and Wetland Processes with Emphasis on Modeling, Marine Science Series , Vol. 11, Plenum Press, New York, NY, 1980.

(2) hardy, Thomas A., and Crawford, Peter L.

,

"Frequency of Coastal Flooding at Roughans

Point, Broad Sound, Lynn Harbor and the Saugus-Pines River System, " Technical Report

in publication, U.S. Army Waterways Experiment

Station, CE, Vicksburg, MS.

(3) Hardy, Thomas A., Prater, Mark D. , and Butler, H. Lee, "Fire Island to Montauk Point, New York, Storm Surge Model Study," Technical

Report in publication, U.S. Army Waterways Experiment Station, CE, Vicksburg, MS.

(4) Prater, Mark D. , Hardy, Thomas A., Butler, H. Lee, and Borgman, Leon E.

,

"Estimating Error of Coastal Stage Frequency

Curves," 19th International Conference on Coastal Engineering, ASCE, Houston, TX, 3-7 Sept, 1984.

(5) Reid, R.O. , and Bodine, B.R. , "Numerical

Model for Storm Surges in Galveston Bay," Journal of Waterways and Harbors Division , ASCE, Vol. 94, No. WWl, Proc. Paper 5805,

Feb 1968, pp. 33-57.

(6) Welander, P., "Numerical Prediction of Storm Surges," Advances in Geophysics , Vol. 8, Academic Press, New York, NY, 1961.

472
Task Committee Reports

Report of Task Committee on

(A) STRONG-MOTION INSTRUMENT ARRAYS AND DATA

Date ; May 13, 1986 Place ; National Bureau of Standards

Gaithersburg, MD USA Attendees ; Japan Side - H. Tsuchida, Chairman (PHRI)

K. Ohtani (NRCDP)

U. S. Side - A. G. Brady, Chairman (USGS)

H. Meyers (NOAA)

I . Activities and Principal Accomplishments to Date Catalogs of strong-motion earthquake records observed both in the United States and Japan are being exchanged. U.S. data are published in the "Seismic Engineering Program Report," and the Japanese data are published in "Strong-Motion Earthquake Records

in Japan."

Discussions were held on the possibility and the necessary

support for accepting a team of scientists and engineers,

together with their instruments, immediately following a large

earthquake in the host country (see No. 10, following).

II . Future Programs

Both sides of the Task Committee plan to continue their respective

efforts in the following ten points, 1 to 10, and to initiate three

points, 11, 12, and 13.

1. After an earthquake which has caused damage

to structures or an earthquake during which

maximum acceleration exceeding about O.lg has been recorded, the task committee of the country where the earthquake has taken place

will provide a list of the strong-motion

earthquake records for the counterpart of the task committee. The list contains maximum

component acceleration of each record. If there is such a list compiled by any organization, the list mentioned above may be replaced

by it.

2. Every year the task committee will exchange

catalogs of the strong-motion earthquake

records during the previous year. The catalog contains maximum component accelerations

and wave forms of major records. If there is such a catalog compiled by any organization,

the catalog mentioned above may be replaced

by it.

3. The Task Committee makes appropriate arrangements

to provide digitizable copies of records when

they are requested. In addition, arrangements

will be made to provide information on the characteristics of the site and structures at the locations where such records are obtained.

4. When the organizations taking part in the Panel publish reports on strong-motion

earthquake records, these organizations will

475
distribute copies of the reports to the other organizations of the Panel interested in them. The Task Committee exchanges lists of the organizations and their Panel representatives who wish to receive the reports.

5. The Task Committee will continue to exchange

digitized data on all the major strong-motion

earthquake records recovered in both countries, under the conditions mutually acceptable to the agencies involved in acquisition of the data.

6. The Task Committee will exchange technical

information concerning seismometers and accelerometers to be developed, and the deployment

and installation of instruments with emphasis

on bore-hole instrximents , and the design of seismic arrays.

7. The Task Committee plans to assist and cooperate, where possible, in the following

areas

:

a. Assistance and cooperation with

governmental organizations in other seismic areas, in order to promote

high quality strong earthquake

motion observations in all seismically

active areas of the world.

b. Assistance and cooperation with

any international effort to record strong ground motion close to the source of a large magnitude shock.

8* The Task Committee feels the UJNR Panel on Wind and Seismic Effects should play a major

cooperative role in the implementation of relevant parts of the Resolution of the International Workshop on Strong-Motion

Earthquake Instrument Arrays held May 1978,

in Honolulu, Hawaii. The exchange of complete information on all aspects of the program, as it develops particularly in Japan

and the United States, will be carried out in the manner of our standard exchange when

appropriate. Particularly with unique data

from arrays, or deep bore-hole instruments in the U.S. and Japan, Task Committee Chairmen

will retain the responsibility of relaying

information on their existence and subsequently

of transferring the digital data from such institutions as, in the U.S., the USGS, CDMG,

use, and UCSD; and, in Japan, PWRI , BRI , PHRI

,

MRI, and NRCDP.

9. The Task Committee recommends collaborative

studies involving strong-motion instrumentation

and interpretation of resulting strong-motion

data sets for purposes of earthquake hazard

reduction.

10. The Task Committee will study the possibility

of exchange of scientists, engineers, and appropriate equipment in a timely fashion for the observations following large earthquakes

in either country. As the largest and most interesting aftershocks occur immediately

476
following the main earthquake, recovery of the most useful aftershock data requires that

the response to major earthquakes be initiated

as rapidly as logistics permit.

11. The task committee will hold a workshop on Strong-Motion Earthquake Observation in the San Francisco area in August 1987. The details of the workshop program will be dis- cussed between both Task-Committee chairmen

and include array and network planning,

recording instrumentation, digitizing hardware,

and the digitization of representative records.

12. The Task Committee recognizes that another workshop on Processing of Seismic Strong

Motion Records to be held in Vancouver in August 1987 is significantly relevant and would complement the Panel's Workshop. The Task Committee urges that the Panel members contribute to this workshop.

13. The Panel discussed the occasional, but extended,

unavailability of existing strong-motion data of engineering

significance from very large earthquakes outside the US and

Japan. The Task Committee considered dealing appropriately

with US and Japanese data. The Task Committee chairmen will extend their committee's study to encourage the appropriate

dissemination of world-wide data having engineering significance

and prepare a report ‘for the next Panel meeting.

477
Report of Task Committee on

(B) LARGE-SCALE TESTING PROGRAM

Date : May 13, 1986 Place: National Bureau of Standards,

Gaithersburg, MD, U.S.A. Attendees : Japan Side - K. Ohtani, Chairman (NRCDP)

U. S. Side - E.V. Leyendecker, Chairman (NBS)

J. J. Burns (NRC) J.B. Scalzi (NSF)

I » Activities and Principal Accomplishments to Date

The main activities related to this task committee during the past year are as follows:

1. The UJNR Panel on Wind and Seismic Effects endorsed the Implementing Arrangement for Cooperation in the U.S. -Japan Earthquake

Research Program Involving Large-Scale

Experiments of Masonry Structures during the 17th meeting in Tsukuba.

2. The Joint Technical Coordinating Committee on

the Masonry Research (JTCCMAR) was held in Tokyo

in August 1985. Coordinated research on masonry structures were discussed.

3. Support tests on masonry research continued

on the Japanese side and were initiated on the U.S. side.

4. Active exchanges of U.S. and Japanese researchers have occurred under the program

and have been very useful.

5. Testing of the full-scale steel structure and support tests have been completed in both countries. Analytical studies are currently

in progress.

II . Future Programs

The status of the U.S. -Japan Joint Program of reinforced concrete, steel and masonry structures were reviewed. The Task Committee

adopted the following resolutions as a result of discussion.

1- The Panel of each country should collect

copies of reports for all coordinated research conducted in that country in a single set. Each country should exchange a set of these reports.

2. The major aspects and conclusions of the steel test programs in each country should be synthesized into a single report. The dissemination activities for accomplishments

of the U.S. -Japan cooperative research program on steel and reinforced concrete structures should be encouraged and supported

by each side.

3. To assure effective exchange of information

on masonry structures, continuing and

478
expanded personnel exchange is encouraged

under the coordinated research program.

4. The Task Committee will review progress of the Joint Research Program and continue to reassess the need for testing of other types

of structures and report its findings and recommendations to the Panel.

5. The Task Committee recommends that the Panel endorse the Coordinated Research Program on Large-Scale Testing of Masonry Structures.

6. The 2nd meeting 'of the Joint Technical

Coordinating Committee on Masonry Research is scheduled for the fall of 1986 in Denver, Colorado.

7. Construction of a full-scale five-story masonry

building will begin in the BRI laboratory in Tsukuba in Japanese fiscal year 1986.

8. Task Committees (B), Large-Scale Testing

Program and Task Committee (D), Evaluation of Performance of Structures should maintain

closer liaison to facilitate the planning of instrumentation and analysis for full-scale

test structures.

9. Technical exchange of information on masonry

structures should continue to be exchanged

through this Task Committee.

10. Exchange of information on large-scale testing

facilities and large-scale testing programs should

be encouraged.

479
Report of Task Committee

(C) TASK COMMITTEE ON REPAIR AND RETROFIT OF EXISTING STRUCTURES

Date ; May 14, 1986 Place; National Bureau of Standards

Gaithersburg, MD, U.S.A. Attendees ; Japan Side - H. Yamanouchi (BRI) U.S. Side - J. B. Scalzi, Chairman (NSF)

A. G. Franklin (WES)

R. D. Hanson (MICH)

R. McConnell (VA)

A. J. Eggenberger (NSF)

I

.

Activities and Accomplishments to Date

In Japan the following activities have been carried out;

1. The Japan National Project - Post - Earthquake Measures of Buildings and Structures Damaged

by Earthquake has been completed as a five-year plan from 1981 to 1985. Research products are summarized as "Manuals

for Repair Methods for Damaged Structures and Buildings"

.

2. "Total System for Aseismic Design, Inspection

and Repair and Retrofit of New or Existing

Building. Equipment and Non-Structural

Element for Official Use" was compiled by

' Government Buildings Department, Ministry of Construction in March, 1985 and is still under trial use. In the U.S. the following activities have been carried out;

1. Studies on base isolation and supplemental

damping devices have been in progress which may be of benefit in repair and retrofit of existing structures.

2. Research has been initiated on techniques to assess current material and mechanical

properties in structures which are aging or showing the effects of weathering or excessive

use. Such information is essential in developing retrofit measures.

3. Research on methods to assess the performance

of existing and damaged structures is continuing along with research on methods to repair and retrofit such structures.

II . Future Programs

1. Continue and expand the exchange of reports, books and special publications related to repair and retrofit of existing structures.

2. A cooperative workshop on repair and retrofit should be held in the near future. Possible

date might be in conjunction with the May

1987 UJNR meeting to be held in Japan.

480
3. Exchange information and give consideration

to possible cooperative efforts to study methods to quickly and economically assess

the characteristics and residual structural capabilities of material which may have suffered degradation due to weathering,

aging, misuse or change of use and on methods

to upgrade the capabilities of such materials

where required. Develop methods of analysis

for predicting the structural capacities of structures containing such materials.

4. Share information such as handbooks and on specific completed projects such as evaluations

of particular designs for which retrofit

studies have been carried out and which may

provide specific validation of retrofit

design methods.

5. Arrange for the evaluation of the reports and manuals on seismic retrofit of concrete

structures of building and contents and of historical structures which were prepared

under the UNIDO Balkans Project.

6. Carry out research to evaluate the potential

structural benefits and relative economics of using base isolation and supplemental damping

devices for retrofitting existing structures. Such devices may be of particular interest

for historical or monumental structures.

7. Devote increased attention to developing

methods to assess nonstructural items in existing structures which could pose severe hazards and/or cause loss of service. Develop methods for retrofit or repair of important

non-structural elements.

8. Cooperate with Committee D "Evaluation of Performance of Structures" to identify and encourage coordinated U.S. -Japan papers on computer programs which are useful for the evaluation of existing structures and which

could be used to guide repair and retrofit

decisions

.

9. Join with Committee H to recommend that the Panel endorse the initiation of a coordinated

research program on remedial measures to mitigate or prevent damage due to liquefaction

failures.

481
Report of Task Cominittee on

(D) Evaluation of Structural Performance

Date : May 13, 1986 Place; National Bureau of Standards Gaithersburg, MD, USA

U. S. Side: G. R. Fuller, Chairman (HUD)

P. E. Gurvin ( STATE

)

J. Z. Mirski (VA)

J. B. Scalzi (NSF)

R. J. Smith (ARMY)

Japan Side; H. Yamanouchi, Chairman (BRI)

H. Okada (BRI)

I . Activities and Principal Accomplishments to Date

In responses to the need for better understanding of current safety evaluation methodologies. Task Committee D organized three

U.S. -Japan workshops on the evaluation of the performance of existing buildings for resistance to earthquakes.

The First Workshop, held in Japan in 1983, treated five theme

areas: (1) Overview of evaluation of buildings for seismic

resistance, (2) Practical methodology, (3) Computer programming,

(4) Evaluation of structural performance through full scale

tests, and (5) Post-earthquake evaluations. The Second Workshop

was held in the U.S. in 1984 on (1) Safety evaluation methods,

( 2 ) Evaluation methods applied to benchmark structures , and ( 3

)

Damage and analysis of other structures. The Third Workshop was held in Japan in 1985 and concentrated on additional comparisons

of evaluations of benchmark structures and on updating information

on methodologies.

Those three workshops, participated in by over 30 from the

U.S. side and over 30 from the Japanese side, were very successful

The detailed methods reported on ranged from rapid classifications

of groups of buildings to detailed dynamic analysis of individual

structures. Methods developed in Japan and in the U.S. were "tuned" to the characteristics of typical buildings in each country. Comparative

analyses of benchmark structures reflected that Japanese seismic

design approaches are relatively more conservative.

The comparisons also reveal that the various methods are suitably

accurate for the expected level of evaluation for buildings with

good arrangement of lateral strength elements, and that a general

idea of expected damage level can be estimated. However,

additional developments are required to validate and improve the methods to make them more useful, especially for buildings of unusual geometry or with special secondary elements, and to include effects of soil-structure interaction. Three volumes of proceedings of the First to Third workshop are scheduled to be published in 1986 by Cornell University, under a grant from the U.S. National Science Foundation.

The U.S. Chairman distributed copies of a paper by Gergely, White, and Fuller on "Seismic Performance of Existing Buildings" to be presented at the U.S. Earthquake Engineering Conference in Charleston, S.C. in August 1986. This paper summarizes the results of the three U.S. Japan Workshops, and discusses the methodologies used to evaluate six benchmark structures.

482
The Task Committee then discussed the status of various related

projects, and objectives and possible future activities to be pursued. It was recommended that each side investigate the possibility of obtaining useful information on evaluating

performance of buildings after the 1985 Mexican Earthquake.

II . Future Programs

The Task Committee reconfirmed the objectives and recommendations

of the May 17, 1984 meeting concerning analysis, instrumentation,

and postearthquake evaluation of benchmark structures; and agreed to continue activities of the Task Committee on the following

tasks

:

1. Encourage the continued exchange of literature,

analyses, and research reports relative to assessment of seismic resistance of buildings.

2. Expand the inventory of existing buildings in each country that have been evaluated and instrumented, and then observed after any future seismic activity. Continue to evaluate other buildings in each country

using Japanese and American methods of analysis.

3 . Develop recommended evaluation and screening

methods for determination of seismic resistance

of existing buildings and identify research

needs to improve evaluation technology.

4. Concentrate on the evaluation of several structural steel buildings in each country

using Japanese and U.S. methodologies. These buildings will be from six to ten stories in height

.

5 Propose conducting a Fourth Workshop on "Evaluation of Seismic Resistance of Existing

Buildings" to be held in Japan prior to the 19th UJNR Meeting. The main objective of this workshop will be to present conclusions

and recommendations for methodologies to analyze seismic strength of buildings.

483
Report of Task Committee on

(E) NATURAL HAZARD ASSESSMENT AND MITIGATION THROUGH LAND USE PROGRAMS

Date ; May 14, 1986 Place ; National Bureau of Standards,

Gaithersburg, MD, U.S.A. Attendees ; Japan Side - M. Nagaoka,* Temporary Chairman (GSI)

U. S. Side - S. T. Algermissen, Chairman (USGS) A.G. Brady (USGS)

I

.

Activities and Principal Accomplishments to Date

1. The US-Japan Workshop on "Urban Earthquake

Hazards Reduction," sponsored by the U.S. Earth- quake Engineering Research Institute ( EERI

)

was held at Stanford University in July

1984. Attendees included UJNR members Dr.Kazuhiko Kawashima and Dr. John B. Scalzi, and representatives of the Government of Shizuoka Prefecture: Messrs. Toshiroh

Sugiyama and Katsunori Imado. The final

report from EERI was published in July 1985 and was distributed to Task Committee (E) members

.

2. The Geographical Survey Institute (GSI) of Japan

has conducted thorough reviews of research

and work on the effects of topographical and geological conditions on seismic damage. The Public Works Research Institute ( PWRI ) of Japan has proposed a procedure for assessing

direct and indirect losses from earthquake

disasters, and the direct and indirect losses were evaluated for the Miyagi-ken-oki earthquake

of 1978 and the Nihon-kai chubu earthquake of 1983. The Building Research Institute (BRI) of Japan

has conducted the development of automatic

computer programs for analyzing the expected

ground intensity at an arbitrary site, and has made analyses of the response characteristics

of ground having an irregular surface and heterogeneity.

3. The U.S. Geological Survey conducts a broad based program in natural hazards reduction. This program includes assessment of; (1) earthquake

ground motion; (2) soil liquefaction potential,

(3) landslide potential; and (4) earthquake

economic losses.

4. Several papers pertaining to Task Committee (E) were presented at the 18th Joint Panel Meeting:

a. Sessions 2 and 4 "Earthquake

Engineering"

b. Session 5 "Storm Surge and Tsunami"

c. Session 7 "Technical Cooperation

related to the Mexican Earthquake

and Chilean Earthquakes"

484
5 . A discussion was held concerning the possibility

of revising the name of the Task Committee.

It was pointed out that the term "Land Use Programs" has a rather narrow meaning in English. The Task Committee recommends that

the Panel endorse the change of name of the Task Committee to "Natural Hazard Assessment

and Mitigation through Land Use Programs".

II . Future Programs

1. Both sides will continue to exchange technical

information on the following subjects:

a. Evaluation of, and mapping procedures

for seismic risk.

b. Maps of geological conditions,

soil liquefaction potential,

seismic risk, and tsunami risk.

c. Seismic response characteristics of ground having an irregular surface, and heterogeneity.

d. Planning methodologies for earthquake

disaster mitigation’ in large

populated cities.

e. Socio-economic aspects of earthquake

disasters and evaluation procedures

for them.

f. Direct and indirect losses from earthquakes and evaluation procedures

for them.

2. Both sides will continue to promote coordinated

research and exchange of specialists in natural hazard assessment and mitigation

through land use programs.

3- The U.S. Chairman will furnish to Task Committee members future reports on natural hazard assessment and mitigation through land use programs being coordinated or sponsored

by the Federal Emergency Management Agency

(FEMA), the Interagency Coordinating Committee

of the National Earthquake Hazards Reduction

Program (ICC/NEHRP), the U.S. Geological

Survey (USGS), the National Science Foundation

(NSF), and the National Academies of Science

and Engineering.

4. The possibility of holding workshops on (1) techniques of earthquake loss estimation; and

( 2 ) evaluation of seismic risk and mapping

procedures, was discussed. It was agreed that further consideration of these workshops would be undertaken during the coming year.

485
Report of Task Committee on

(F) DISASTER PREVENTION METHODS FOR LIFELINE SYSTEMS

Date ; May 14, 1986 Place: National Bureau of Standards,

Gaithersburg, MD, U.S.A

Attendees: U.S. Side - S. C. Liu, Chairman (NSF) H.S. Lew (NBS)

C. Scheffey (DOT) Japan Side - Y. Sasaki, Temporary Chairman (PWRI)

H. Tada (PWRI)

I . Activities and Principal Accomplishments to Date

A US-Japan Workshop on Seismic behavior of Buried Pipelines and Telecommunication Systems was held in December at the Public Works Research Institute, Japan. The proceedings of this workshop

were published by Columbia University and were distributed to the Workshop participants, appropriate members of the UJNR Panel on Wind and Seismic Effects, and other interested parties. In Japan, the Ministry of Construction has completed a 5-year

(1981-1985) research project on the development of repair

technology for seismically damaged structures. As a result of this project, a Manual of Repair Methods for Civil Engineering

Structures Damaged by Earthquakes (Draft) was prepared by PWRI. In the Manual procedures of damage surveys and repair work for buried sewer pipelines are described. An outline of the results

of this research project was introduced at the 18th UJNR Joint

Meeting of the Panel on Wind and Seismic Effects. PWRI is also conducting a dense instrument array measurement program at PWRI campus and in Shizuoka Prefecture to identify ground motions

applicable to seismic design of buried pipeline structures. An outline of the dense instrument array program was presented at the 18th UJNR Joint Meeting. The Ministry of Construction is promoting the construction of cable box systems ( CAB systems

)

under urban highways, in order to install aerial cables (for

electricity and telecommunication) into underground cable boxes, and has prepared Technical Standards for CAB Systems (Draft). In conjunction with these activities, PWRI has proposed a seismic

design procedure of CAB systems and conducted dynamic experiments

for examining the seismic safety of CAB systems. An outline of the Technical Standards for CAB Systems (Draft) and the dynamic

experiments was reported during the 18th UJNR Joint Meeting.

In U.S., a study program to assess the design and operating

standards for lifeline systems has been initiated by the Building

Seismic Safety Council (BSSC) under the support of the Federal Emergency Management Agency ( FEMA )

.

The American Society of Civil Engineers (ASCE) through its Technical Council on Lifeline Earthquake Engineering (TCLEE) and

the American Society of Mechanical Engineers (ASME) have developed

and published several design manuals and advisory notes dealing

with the seismic design of lifeline systems. These efforts are made with contributions from investigators supported by the National Science Foundation (NSF). NSF has continued the support of a number of research projects

dealing with lifeline earthquake engineering. For example, there

is an International Symposium/Workshop between U.S., Japan, and Taiwan on lifeline system in 1985, and a research project on telecommunication systems and another project on field experiments

of buried pipelines. The workshop proceedings and the project

reports have been exchanged through the UJNR channels.

486
The strong motion array project in Taiwan (SMART) has been continuously operated and new data has been recorded. In 1985 NSF supported a new project in conjunction with the SMART array in deploying a instrumentation system to measure the soil responses during the earthquakes. The data recorded from these projects

will be continuously exchanged between the two countries.

II . Future Programs

1. The members of the Task Committee will

facilitate the continued exchange for technical information and cooperation in research on specialized problems listed as follows

:

a. Seismic damages to lifeline systems

b. Seismic design procedures for lifeline systems

c. Procedures and instriomentation to detect and inspect damage and the behavior of lifeline structures including buried pipelines.

d. Repair and retrofit methodology for lifeline systems

e. Estimation of reliability of lifelines

f. Management and public education on importance of lifeline systems

g. Investigation of needs for large-scale

testing of lifeline systems.

2. It is recommended to initiate a US-Japan

coordinated research project on seismic

behavior, seismic design method, and seismic

reliability of buried pipelines and similar

lifeline systems. In order to make up a .specific program of the future US-Japan

coordinated research project including a possible workshop on Lifeline Earthquake

Engineering, it is also recommended to hold a planning meeting of the Task Committee during

the Charleston conference in August 1986.

3. The Third U.S. National Conference on Earthquake Engineering will be held in Charleston, South Carolina in August, 1986. Since a special session on lifeline earthquake

engineering is being scheduled, it is proposed that Task Committee (F) members participate in the special session.

4. The Task Committee recognizes the importance

and potential impact to the engineering

practice of the Manual of Repair Methods for Civil Engineering Structures Damaged by

Earthquakes (Draft and in Japanese) prepared

by PWRI, and recommends that consideration be given to translate this document into English

for wide distribution to the U.S. earthquake

community.

487
Report of Task Committee on

(G) WIND CHARACTERISTICS AND STRUCTURAL RESPONSE

Date ; May 14, 1986 Place: National Bureau of Standards,

Gaithersburg, MD, U.S.A. Attendees: Japan Side - H. Okada, Chairman (BRI) U.S. Side - R. D. Marshall, Chairman (NBS)

I o Activities and Principal Accomplishments to Date Noting that:

a. High winds frequently cause loss of life and extensive property damage.

b. The knowledge derived from the exchange of high wind data and information on wind effects can be useful in reducing loss of life and property damage,

c. Insufficient knowledge exists on the effects of high winds on structures and methods for modeling

these effects, and

d. The needs for wind data for various

applications and new projects are increasing.

The Task Committee carried out the following activities;

1. The following documents were exchanged:

a. Rcjport entitled "Hurricane Alicia - Glaveston and Houston, Texas, August 17-18, 1983," published by

Committee on Natural Disasters,

Commission on Engineering and Technical Systems, National Research Council, USA.

b. Report entitled "Performance Requirements and Preliminary Design

of a Boundary Layer Wind Tunnel

Facility," NBS Report No. NBSIR85-3168

,

May 1985.

c. Design and construction details of Japanese boundary layer wind

tunnels were provided during the past year. This information is in addition to data on Japanese wind

tunnels published in 1984.

2. Some instruments capable of measuring high

winds at coastal or sea regions were installed

or prepared in Japan: Kashima coast observation

with several masts 20 to 30 m high, an off-shore tower 20 m high at Hiratsuka, wind measurements at long span bridges in the Setonaikai Sea, et al.

3. Comprehensive measurements of surface pressures and internal pressures acting on a full-scale low-rise building situated in open

488
terrain. This proposed three-year study is under consideration by the National Science

Foundation, USA.

4. Development work was initiated by BRI for a research effort involving wind pressures acting on high-rise and on low-rise buildings.

A priority area for high-rise buildings is the reliable measurement of internal pressures

in buildings with large open interiors. The motivation for additional studies of low-rise

buildings is to improve on the quality of measured surface pressures' such as those reported in the Aylesbury study in England.

5. Concerning high wind effects, particular

investigations continued in the U.S.A. on various projects, such as wind loading and strength prediction of glass cladding and wind loading and response of transmission towers.

II . Future Programs

1. Exchange observation reports of high winds.

2. Encourage interaction between meteorologists

and engineers to identify the types of wind

data required for future use in establishing

extreme wind distributions; in determining

wind loading on buildings and structures; in understanding the urban wind climate; and in considering design issues involved with the wind generation of energy.

3 . Encourage the establishment of minimum

requirements for the simulation of atmospheric

boundary layers in wind tunnels and the exchange of boundary layer wind tunnel test results, including comparisons with corresponding

full-scale situations.

4. Exchange information on criteria, techniques

and instrumentation for structural modeling

in boundary layer wind tunnels. Also encourage the exchange of information of measurements made on actual structures and on comparisons with predictions from wind tunnel studies of these structures.

5. Expand the exchange of engineers and meteorologists, and encourage the mutual use of available research facilities in the two countries.

489
Report of Task Committee on

(H) Soil Behavior and Stability During Earthquakes

Date : May 13, 1986 Place ; National Bureau of Standards Gaithersburg, MD, USA Attendees : U.S. Side -.A. G. Franklin, Chairman (WES)

F. G. McLean (USER)

R. L. Dewey (USER)

Japan Side - Y. Sasaki, Chairman (PWRI)

M.Nagaoka (GSI)

I . Activities and Accomplishments to Date

1. Exchange of relevant documents on soil behavior during earthquakes was made. From

the Japanese side, eight papers were submitted

and from the U.S. side, three papers were submitted to the Eighteenth Joint Meeting of the Panel.

2. Following the resolution of the Seventeenth

Joint Meeting, a workshop on In-Situ Testing

for Evaluation of Soil Liquefaction Susceptibility

was held in San Francisco, California, in August,

1985. There were 10 representatives from the Japanese side and 14 from the U.S. side.

3. Under the auspices of the UJNR, Mr. Takashi

Tsuchida of the PHRI is being hosted by WES

as a visiting engineer, working on liquefaction

of fine-grained soils and dynamic behavior of embankment dams. The term of this visit is for one year beginning in November 1985.

4. Under the auspices of the UJNR, Dr. Nobuyuki

Narita of the PWRI visited the geotechnical

centrifuge at the University of Colorado and the Engineering and Research Center of the USER. He discussed with the USER the visit to the PWRI of

Mr. Jeffrey Farrar of the USER as an exchange

researcher hosted by the Japanese government.

II . Future Programs

1. Proceedings of the 1985 workshop on In-Situ

Testing for Evaluation of Soil Liquefaction

Susceptibility will be published.

2. Mr. Jeffrey A. Farrar will visit the PWRI in October and November 1986.

3. The committee recommends that the panel endorse the initiation of a cooperative research program on remedial measures to mitigate or prevent damage due to liquefaction

failures

.

4. When requested, the Task Committee will

assist in arrangements for visits to earth embankments including fill dams and foundations

490
which have been subjected to significant

ground motions.

5 . The Task Committee encourages the maximum

exchange of information on the following

areas of earthquake engineering which are currently of intense interest to both

U.S. and Japanese engineers.

a. Liquefaction potential of soils

b. In-situ testing methods for evaluation of soil liquefaction

potential

c. Seismic stability of embankment

structures and natural slopes

d. Remedial treatment of embankments

and foundations to improve seismic

stability

e. Seismically induced permanent

displacements in earth structures

f. Seismically induced settlements of soils and foundations

g. Earthquake effects on semi-buried concrete structures

h. Earthquake effects on retaining walls

i. Centrifuge testing facilities and methods

j . Laboratory testing of models of geotechnical structures

k. Field performance data showing full-scale seismic response of earth dams

6. The Task Committee requests that the Panel encourage

visits and exchanges of engineers and scientists between U.S. and Japanese government research organizations,

and that the Panel assist such exchanges by providing

the channels of communication for making the arrangements,

and endorsing such visits and exchanges as UJNR

activities, when appropriate.

491
Report of Task Cominittee on

(I) STORM SURGE AND TSUNAMI

Date : May 14, 1986 Place ; National Bureau of Standards,

Gaithersburg, MD, U.S.A. Attendees : U.S. Side - C. S. Barrientos, Ghairman (NOAA)

H. Meyers (NOAA)

C. Astill (NSF)

R. L. Dewey (USBR)

Japan Side - K. Ohtani, Temporary Chairman, (NRCDP)

H. Tsuchida (PHRI)

I . Activities and Principal Accomplishments to Date

1. The U.S.side (NOAA) supplied the Japan side

(MRI) with digital records of tsunamis affecting

Japan. MRI has examined the records and identify

the differences with their data. The U.S. side

(NOAA) will make the necessary adjustments to conform with the Japanese information.

2. The Japan side (PWRI) has sent copies of photographs taken right after the MihonkaiChubu Earthquake tsunami to the U.S. (NOAA).

These photographs are important information in understanding tsunami effects.

3. The U.S. side (NOAA) has provided the Japan

side (MRI) with the Pacific Ocean Bathymetry

Data Base. The data base is used in modeling

tsunamis and in producing tsunami propagation

charts. Based on the bathymetry data, three charts were produced by NOAA and the charts were given to the Japanese members during the 18th Panel meeting.

4. A report by NOAA (SE-39) on tsunamis in Peru-Chile

was given to the Japan side during the 18th meeting.

II . Future Programs

1. The Task Committee supports the continuation

of exchange of data and information on tsunami and storm surge. The Committee also supports exchange of researchers whenever

possible.

2. The Task Committee proposes to hold a Second

Tsunami and Storm Surge Workshop in connection

with the 19th Panel Meeting in Japan. Details

of the workshop will be arranged by the Task

Committee Chairmen with the approval of the Panel Chairmen. Specifically the workshop

will address computational and experimental

methods in tsunami and storm surge. The

U.S. National Science Foundation will

consider providing partial support for the

U.S. side to attend the Workshop.

492
This proposed workshop is subject to the final approval of the Japan Task Committee members.

3. The U.S.side (NOAA) and the Japan side (MRI) will work together on tsunami propagation

studies which will include producing tsunami

propagation charts.

4. The U.S. side (NOAA) will give to the Japan

side the report and documentation of the Storm

Surge Model, SLOSH (Sea, Land and Overland

Surges from Hurricanes).

493
Report of Task Committee on

(J) WIND AND EARTHQUAKE ENGINEERING FOR TRANSPORTATION SYSTEMS

Date ; May 13, 1986 Place ; National Bureau of Standards,

Gaithersburg, Maryland, U.S.A. Attendees ; Japan Side - H. Tada, Chairman ( PWRI

)

U.S. Side - C. F. Scheffey, Chairman (FHWA)

H. S. Lew (NBS)

J. B. Scalzi (NSF)

I

.

Activities and Principal Accomplishments to Date The major accomplishments of Task Committee (J) during the past year were as follows:

1. Summaries of wind and earthquake research on bridges conducted by both countries were presented to the 18th Joint Meeting, and exchange of information was made.

2. The coordinated research study on the seismic

performance of bridge piers and columns was begun in 1983, and the following experiments were conducted.

a. performance of reinforced concrete

piers and columns subjected to dynamic cyclic loading

b. model tests on the failure of reinforced concrete piers

c. testing full-scale concrete columns

and associated small-scale model

tests

d. behavior of concrete filled steel

tubes

3. The report which consists of a compilation of bridge research papers from recent UJNR Meeting was published and sent to the Japanese Task Committee in June, 1985.

4. The Second U.S. -Japan Bridge Workshop was held in San Francisco, California, in August

19-22, 1985 with cooperation from organizations

concerned in both countries. Thirteen

U.S. and twelve Japanese members participated

in the workshop. Exchange of technical

information was made and comprehensive

discussion were held on future cooperative

programs. The proceedings made by U.S. side were distributed to the participants of the both countries in December 1985. Some of the subjects are:

a. Research on aerodynamic stability

of suspension bridges and cable-stayed

bridges

b. Laboratory experiments on the seismic behavior of reinforced

concrete piers and columns including

concrete-filled steel tubes

494
c. Repair procedure for seismically

damaged bridge structures

d. Experimental field testing of bridges including forced and ambient vibration measurements

e. Experimental and analytical studies

on the effects of soils on the behavior of bridge structures

f. Strengthening procedures for existing bridges

g. Strong motion instrumentation of bridges, corresponding data processing and parameter identification

h. Seismic design and construction

procedures of highway bridges

5. Dr. H. Tada, Chairman of Task Committee (J), stayed in the U.S. for one month beginning in August 9, 1985 and visited laboratories and organizations related to the Panel on the basis of personnel exchange program of the UJNR Panel.

II . Future Programs

As a result of comprehensive discussions, the Task Committee (J) hereby resolves to carry forth with the following programs:

1. Continue with the coordinated experimental research study on the seismic performance of bridge piers and columns, and review the progress of the study at the intermediate

stage.

2. Develop a coordinated research study, both theoretical and experimental, on the seismic,

aeroelastic, and aerodynamic response of cable-supported bridges. A specific proposal

for reciprocal testing of models in turbulent

flow is to be developed by the co-chairman.

3. Encourage the continued exchange of information

and research personnel through the Panel on Wind and Seismic Effects of UJNR.

4. Mr. Koichi Minosaku, researcher of Bridge

Division, PWRI, MOC, will be a guest researcher

at the Center for Building Technology, NBS,

U.S. A. for 10 months beginning in September,

1986, to perform cooperative experimental

research on the seismic performance of bridge

piers and columns. This assignment is part

of the personnel exchange program of the UJNR

Panel.

5. The Task Committee endorses the initiation of a coordinated and expanded program of research on remedial measures to mitigate or prevent damage to existing bridges

by the natural hazards of earthquakes and strong winds.

6. The Task Committee proposes the third Bridge Workshop

to promote cooperative research on wind and seismic-resistant design of long span bridges and strengthening of existing bridges. The Workshop will

be held in Japan in conjunction with the 19th joint meeting of the Panel or in October, 1987, in conjunction with the completion of Honshu Shikoku

Bridges, Ko j ima-sakaide Route. Details of the Workshop should be determined based on discussion

between the Task Committee Chairmen.

495
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NBS-114A (REV. 2 -ac) U.S. DEPT. OF COMM. BIBLIOGRAPHIC DATA

SHEET (See instructions)

4. TITLE AND SUBTITLE

1. PUBLICATION OR REPORT NO. NBSIR 87-3540

2. Performing Organ. Report No, 3. Publication Date April 1987 Wind and Seismic Effects - Proceedings of the 18th Joint UJNR Panel Conference

5. AUTHOR(S) Noel J. Raufaste, Editor

6. PERFORMING ORGANIZATION (If joint or other than NBS. see instructions) 7. Contract/Grant No. NATIONAL BUREAU OF STANDARDS DEPARTMENT OF COMMERCE 8. Type of Report & Period Covered XMMIfitMIIG[IllMNHMXMZ»®MX Gaithersburg, MD 20899 FINAL

9. SPONSORING ORGANIZATION NAME AND COMPLETE ADDRESS (Street. City, State. ZIP)

10. SUPPLEMENTARY NOTES

I I Document describes a computer program; SF-185, FIPS Software Summary, is attached.

11. ABSTRACT (A 200-word or less factual summary of most significant in formation. If document includes a si gnificant

bi bliography or literature survey, mention it here) The 18th Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects was held

at the National Bureau of Standards, Gaithersburg, Maryland, from May 12 through

May 15, 1986. This publication is the proceedings of the Joint Meeting, and includes

the technical papers and the members of task committees and formal resolutions.

Subjects covered in the papers presented to the panel include: (1) wind engineering,

(2) earthquake engineering, (3) storm surge and tsunamis, (4) U.S. -Japan cooperative

research program, and (5) Mexico City and Chilian Earthquakes.

12. KEY WORDS (Six to twelve entries; alphabetical order; capitalize only proper names; and separate key words by semicolon s) accelerograph; codes; design criteria; disaster; earthquakes; earthquake hazards; geotechnical engineering; ground failure; liquefaction; pipeline; seismicity; solids; standards; structural engineering; structural response; tsunami; wind loads ; winds

13. AVAILABILITY 14. NO. OF PRINTED PAGES

|X)^ Uni imited

1 1 For Official Distribution. Do Not Release to NTIS

r~l Order From Superintendent of Documents, U.S. Government Printing Office, Washington, D.C. 488

15. Price 20402. Order From National Technical Information Service (NTIS), Springfield, VA. 22161 $36,95

SCOMM-DC 6043-P80
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