Wind and Seismic Effects

Proceedings of the 22nd Joint Meeting

NIST SP 796

NATL INST. OF STSND S, TECH R.IC. NIST

PUBLICATIONS , A11103 H37S73

-oc

100

.U57

TMENT OF COMMERCE

#796 itute of Standards and Technology

1990
NATIONAL INSTITUTE OF STANDARDS

TECHNOLOGY

Research MormatkHi Center

Gakh^btirg, MD 20899

DATE DUE

Demco, Inc. 3*-293
Wind and

Seismic

Effects

NIST SP 796

PROCEEDINGS OF

THE 22ND JOINT

MEETING OF

THE U.S.-JAFAN

COOPERATIVE PROGRAM

IN NATURAL RESOURCES

PANEL ON WIND AND

SEISMIC EFFECTS

Issued September 1990

Noel J. Raufaste,

EDITOR

Center for Building Technology

National Engineering Laboratory

National Institute of Standards and Technology

Gaithersburg, MD 20899

U.S. DEPARTMENT OF COMMERCE

Robert A. Mosbacher, Secretary

National Institute of Standards and Technology

John W. Lyons, Director
National Institute of Standards and Technology Special Publication 796

Natl. Inst. Stand. Technol. Spec. Publ. 796, 524 pages (Sept. 1990)

CODEN: NSPIJE2

U.S. GOVERNMENT PRINTING OFFICE

WASfflNGTON: 1990

For sale by the Superintendent of Documents, U.S. Government Printing Office, Washington, DC 20402-9325
PREFACE

Background

Responding to the need for improved engineering and scientific

practices through exchange of technical data and information,

research personnel, and research equipment, the United States

and JAPAN in 19 61 created the US-JAPAN Cooperative Science

Program. Three collateral programs comprise the Cooperative

Science Program. The US-JAPAN Natural Resources Development

Program (UJNR) , one of the three, was created in January

1964. The objective of UJNR is to exchange information on

research results and exchange scientists and engineers in

the area of natural resources for the benefit of both

countries. UJNR is composed of 17 Panels each responsible

for specific technical subjects.

The Panel on Wind and Seismic Effects was established in 1969.

Seventeen U.S. and six Japanese agencies participate with

representatives of private sector organizations, to develop

and exchange technologies aimed at reducing damages from high

winds, earthquakes, storm surge, and tsunamis. This work is

produced through collaboration between U.S. and Japanese member

researchers working in ten task committees. Each committee

focuses on specific technical issues, e.g., earthquake strong

motion data. The Panel provides the vehicle to exchange

technical data and information on design and construction of

civil engineering lifelines, buildings, and water front

structures, and to exchange high wind and seismic measurement

records. Annual meetings alternate between JAPAN and the U.S.

(odd numbered years in JAPAN; even numbered years in the

U.S.). These one-week technical meetings provide the forum

to discuss on-going research and research results; one-week

technical study tours follow the meetings.

The National Institute of Standards and Technology (NIST)

provides the U.S. -side chairman and secretariat. The

Public Works Research Institute (PWRI) , JAPAN, provides the

JAPAN-side chairman and secretariat.

The Panel has organized formal joint research programs. In

1981 cooperative research in Large-scale Testing commenced

under the auspices of the Panel. Also in 1981, joint research

on Reinforced Concrete Structures was initiated. Full-scale

testing was performed at the Building Research Institute (BRI)

,

JAPAN, one of the six Japanese member organizations, with

supporting tests in JAPAN and in the U.S. Two years later

a joint research program on Steel Structures was initiated.

Full-scale testing again was led by BRI with supporting tests

in the U.S. and JAPAN. The US-JAPAN coordinated program for

Masonry Building Research was started in 1985. A coordinated

program for testing lifeline structures resulted in NIST's

testing of full-scale bridge piers in its 50 Mn universal

iii
testing facility to evaluate post-1972 seismic design criteria.

Task Committee meetings, exchanges of data and information

through technical presentations at the annual Panel meetings,

exchanges of guest resea -hers, visi' " to respective research

laboratories and informax interactions Detween Panel meetings,

joint workshops and seminars, and joint cooperative research

programs have contributed to the development and effective

delivery of knowledge that has influenced design and

construction practices in both countries.

Direct communication between counterpart country organizations

is the cornerstone of the Panel. Effective information

exchanges and exchanges of personnel and equipment have

strengthened domestic programs of both countries. There

are opportunities for experts in various technical fields

to get to know their foreign counterparts, conduct informal

exchanges, bring their respective views to the frontiers of

knowledge, and advance knowledge of their specialties.

The Panel's results have supported improvements in practices

in both countries, they have for example:

1. created and exchanged digitized earthquake records

for use as the basis of research for Japanese and

U.S. geotechnics and structures;

2. produced data that advanced U.S. design and

construction of bridge columns;

3. produced large-scale testing data that advanced

seismic design standards for buildings;

4. created a database comparing Japanese and U.S.

standard penetration tests to improve seismic

design criteria for soil liquefaction;

5. created databases on storm surge and shore line

interaction and on tsunamis and tsunami warning

systems for use by designers to verify mathematical

models of tsunamis and storm surge.

22nd Joint Panel Meeting

The 22nd annual U.S. -JAPAN Joint Panel meeting was held at

NIST during 15-18 May 1990. Thirty-seven U.S. members from

17 Federal agencies, universities, and a trade association

participated. JAPAN-side participation included twelve

members from their six member Government agencies and one

from academia. Thirty-nine papers were authored; 21 by U.S.

members and 18 by Japanese. Thirty-two oral presentations

centered on: Wind Engineering; Earthquake Engineering;

Storm Surge and Tsunamis; US-Japan Cooperative Masonry

iv
Research Program; Loma Prieta Earthquake of 17 October 1989;

and reports from just completed Panel Task Committees Workshops

on "Repair, Retrofit, and Performance of Structures" and the

6th Bridge Workshop on "Analysis, Design, Performance, and

Strengthening of Bridges". , Some highlights of the meeting are listed below:

o eight of the Panel's ten Task Committees scheduled

workshops within the coming 12 months;

o a tenth Task Committee was created on Passive, Active,

and Hybrid Control Systems;

c a joint initiative was approved in support of the

International Decade for National Disaster Reduction

(IDNDR) ; work starts with an international workshop

in JAPAN, May 1991; joint U.S. -JAPAN research

will be identified for targeted countries;

o two Japanese reports on base isolation systems were

translated into English for distribution to the U.S.

engineering community.

Refer to Resolutions of the 22nd Joint Meeting in this

section for additional Panel information; refer to Task

Committee Reports, found at the end of these Proceedings,

for specific Panel accomplishments.

During the following week, 19-26 May, Panel members visited

construction projects, disaster areas, and laboratories

including:

Sunshine Skyway Cable Stayed Bridge, Tampa FL. A 12 00 foot

main span, 4.1 mile cast in place $250 million bridge replaced

a twin truss bridge destroyed by a freighter in 1980. This

is one of 12 completed cable-stayed bridges in the U.S. A

set of three shock absorbers serve as effective damping

devices for cable vibration.

Historic Charleston Foundation and The Citadel, Charleston,

SC. A two day workshop focused on effects of Hurricane

Hugo of 21 September 1989 and strengthening and rehabilitating

contemporary and historic structures against multi-hazard

environments. A tour of hurricane damages was made through

historic Charleston and to two barrier islands — Sullivan

and Isle of Pine. Workshop concluded with a TV press

conference.

Harbor (I-llO) Transitway and Bridge Retrofit, Los Angeles

CA. CALTRANS hosted visits to construction of I-llO's 2.6

mile elevated bus/carpool HOV freeway section of the 19.6

mile ($435 million) at-grade highway linking San Pedro to

Santa Monica. At the interchange of Routes 10/210 discussions

focused on CALTRANS retrofit of the elevated bridge ramp to

better resist earthquake loads.

V
Offshore Platform Conference, Long Beach, CA. Four

presentations were made followed by visit to an offshore

platform:

Mineral Management Service, DOI is the regulatory agency

for the offshore oil and gas industry. They lease sea plots (3X3 miles) between the 3 and 200 mile limit (some are

between the 9 and 200 mile limit)

.

Sandia National Laboratory described a new long-life

seafloor earthquake measurement system (SEM) to collect

earthquake data near offshore oil production fields. The

SEM is battery-powered (8 yr. expected life) digital data

acquisition system that telemeters data by a sonar link.

SEM earthquake data indicates the seafloor 's vertical

acceleration is nearly an order of magnitude weaker than

corresponding on-shore vertical motions. The data show that

dryland shakes in vertical and horizontal directions while

seafloors shake predominately in horizontal directions.

These results aid in designing earthquake resistant

offshore structures.

Scripps Oceanoqraphic Institute has a joint project

with the Port and Harbour Research Institute, JAPAN; one of

the six Japanese Government members participating in the

Panel. The joint work involves developing numerical models

for creating computer codes for seismic response analysis

of offshore structures' pile foundations and model tests.

Shell Oil Company leases sea plots off the east,

south, and west U.S. coasts. Platforms in the southern

California area are designed for strength of a 200 year

return period and ductility for a 1000 year return period.

Service-life of structures is 20 years.

The Panel members visited the offshore platform Eureka,

located nine miles off-coast of Long Beach, CA. The

structure is adjacent to the Palos Verde fault. Its

discovery well was drilled in 1976; pumping commenced in

1980. Three platforms are operational in this area; 2-

drilling (Eureka and Ellen) and 1-production (Elly) . Ellen

and Elly are in 265 feet of water (weight, 24K tons)

.

Eureka is about one mile away and is in 700 feet of water.

Its weight is 38K tons. Each drilling platform has

capability to drill 80 wells; about 60 percent of well

drilling capacity has been used. Production of Eureka is

about 14K barrels per day.

UCSD Structural Engineering Laboratory, San Diego, CA.

Discussions centered on research on developing a systems

approach to seismic retrofits of bridge columns. Laboratory

work is heavily supported by CALTRANS . Observed testing a

retrofitted concrete column to earthquake loads. The

procedure involves wrapping a column with prestressed wire.

vi
Testing reveals this procedure to be effective against seismic

loads. Also observed was a test specimen portion of a

full-scale bridge deck which was retrofitted against

earthquakes.

The Panel's efforts are exemplary of effective joint research

and of technology delivery between researchers from JAPAN and

the U.S. Since its conception 22-years ago, about 2000 papers

have been presented, more than 4 0 joint workshops and

conferences have been held, and over 100 guest researchers

have been exchanged. Published proceedings document the

Panel's work. The Panel provides important information

about the U.S. and Japan's civil engineering thrusts which

influence both countries' research and provide the basis

for improvements in building codes and standards.

Lastly, this work was made possible by financial support

from the: Bureau of Reclamation; Corps of Engineers;

Department of State; Federal Emergency Management Agency;

Federal Highway Administration; Minerals Management

Service; National Science Foundation; Naval Facilities

Engineering Command; and Nuclear Regulatory Commission.

Noel J. RAUFASTE, Secretary

U.S. -Side Panel on Wind and Seismic Effects

vii
ABSTRACT

The 22nd Joint Meeting of the U.S. -Japan Panel on Wind and

Seismic Effects was held at the National Institute of

Standards and Technology during May 15-18, 1990. This

publication, the proceedings of the Joint Meeting, includes

the program, list of members, panel resolutions, task

committee reports, and 39 technical papers.

The papers were presented under six themes: (I) - Wind

Engineering, (II) - Storm Surge and Tsunamis, (III) - Joint Cooperative Research Program, (IV) - Earthquake

Engineering, (V) - Loma Prieta Earthquake, and (VI) - Summaries of Task Committee Workshop Reports.

KEYWORDS: accelerograph ; bridges; codes; concrete; design

criteria; disaster; earthquakes; geotechnical engineering;

ground failures; inelastic; lifelines; liquefaction; Loma

Prieta; masonry; repair and retrofit; risk assessment;

seismicity; soils; standards; storm surge; structural

engineering; tsunami; and wind loads.

viii
CONTENTS

PAGE

PREFACE iii ABSTRACT viii AGENDA FOR 22nd JOINT UJNR MEETING xiii

LIST OF PANEL MEMBERS xxi

LIST OF TASK COMMITTEE MEMBERS xxx RESOLUTIONS xxxviii

THEME I: Wind Engineering

Meteorological Data From Hurricane Hugo 3

Joseph H. Golden

Monitoring the Aerodynamic Performance of a Qassic Suspension Bridge 24

Harold R. Bosch

Cooperative Research Program on Wind Tunnel

Testing in Turbulent Flow 33

Koichi Yokoyama, Robert L. Wardlaw, Hiroshi Sato, Takaaki Kusakabe, and Michael G. Savage

Wind on Offshore Platforms 45

Charles E. Smith

THEME II: Storm Surge and Tsunamis

Deep Ocean Recordings of Tsunamis 53

Frank L Gonzalez

Measurement and Modeling of Coastal Storm Effects 59

William E. Roper, H. Lee Butler, and Andrew Garcia

Storm Surge Modeling in the United States 66

Chester P. Jelesnianski, Jye Chen, and Wilson A. Shaffer

THEME III: Joint Cooperative Research Program

The Japan Research Plan on Precast Concrete Seismic Structm^

System (PRESSS Japan) 77

Shin Okamoto, Shinsuke Nakata, Hisahiro Hiraishi, Takashi Kaminosono,

and Hitoshi Shiohara

ix
CONTENTS

PAGE

U.S. Coordinated Program for Masonry Building Research

Fifth Year Status 89

James L. Noland

Design Guidelines of Medium Rise RM Buildings 100

Yutaka Yamazaki, Shin Okamoto, Tsuneo Okada, Akira Matsumura,

and Toshiyuki Kubota

U.S.-Japan Cooperative Research Program on Hybrid

Control of Seismic Response of Bridge Structures 117

Jiro Taguchi, Toshio Iwasaki, Yoshio Adachi, Yasushi Sasaki,

and Kazuhiko Kawashima

THEME IV: Earthquake Engineering

Lateral Resistance of a Pile 131 Michio Okahara, Shoji Takagi, and Keiji Taguchi

Major Seismic Building Upgrades at the Lawrence Livermore National Laboratory 142

F. J. Tokarz, G. E. Freeland, and J. R. Hill

Attenuation Characteristics of Ground Strains Induced During Earthquake 166

Ken-ichi Tokida, Keiichi Tamura, and Koh Aizawa

Dynamic Centrifugal Model Tests of Embankments

on Liquefiable Grounds 178 Yasuyuki Koga and Jun-ichi Koseki

Seismic Resistance Requirements of Roller Compacted Dams 186

Alan T. Richardson and Louis H. Roehm

Damage of Soil Liquefaction Caused by Earthquakes

and Use of Geomorphological Maps 198 Takekazu Akagiri

Dynamic Deformation Characteristics of Compacted

Rockfills by Cyclic Torsional Simple Shear Tests 209

Norihisa Matsumoto, Nario Yasuda, Masahiko Ohokubo,

and Yasushi Kinoshita

Allowable Residual Displacement of Gravity Quaywall

Given by Optimum Seismic Coefficient From

Economical Viewpoint 221

Tatsuo Uwabe

X
CONTENTS

PAGE

The Relation Between the Building Damages and

the Vibration Properties of Ground

(In the Case of Spitak Earthquake) 234

Keiichi Ohtani and Hiroyoshi Kobayashi

New Seismic Design Specifications of Highway Bridges in Japan 253

Toshio Iwasaki, Kazuhiko Kawashima, and Kinji Hasegawa

Scatter on Mechanical Properties of Structiu^

Steels Recently Supplied in Japan 265

Hirofumi Aoki, Hiroyuki Yamanouchi, and Ben Kato

Development of Effective Use of New Metallic Materials for Building Structures 276

Hiroyiiki Yamanouchi, Hirofumi Aoki, and Tatsuo Murota

THEME V: Loma Prieta Earthquake

Initial Overview of the San Francisco Bay and

Santa Cruz Mountains Ground Motion 283 A. Gerald Brady

The Loma Prieta Earthquake:

Site Response and Liquefaction. 289

Ellis L. Krinitzsky and A. G. Franklin

Dam Performance During the Loma Prieta Earthquake of October 17, 1989 306

Robert B. Mac Donald

Performance of Wood-Framed Structures

in the Loma Prieta Earthquake 324

John Tisseli

Seismic Performance of Nonstructural Elements

During the Loma Prieta Earthquake 331

T. T. Soong

Socio-Economic Impacts of Lifeline Performance

Related to the Loma Prieta Earthquake 337

Maryann T. Phipps and Ronald T. Eguchl

xi
CONTENTS

PAGE

Lifelines Performance During the October 17, 1989

Loma Prieta Earthquake 344

M. Khater, C. Scawthorn, J. Isenberg, L. Lund,

T. Larsen, and M. Shinozuka

Summary Report on the Loma Prieta Earthquake

of October 17, 1989 363

Toshio Iwasaki

Manuscripts Authored for Panel Meeing but Not Presented Orally

Experimental Study on Lateral Flow of Ground

Induced by Soil Liquefaction 377

Yasushi Sasaki, Ken-ichi Tokida, Hideo Matsixmoto, and Shoichi Saya

Development of Advanced Reinforced Concrete Buildings

Using High-Strength Concrete and Reinforcement 388 Tatsuo Murota, Hisahiro Hiraishi, Takashi Kaminosono,

Masaomi Teshigawara, and Hitoshi Shiohara

Characteristics of Supplemental Dampers for Earthquakes 400

Robert D. Hanson

An Economic Analysis of Laminated Rubber and

Steel Bearing Base Isolation System for Buildings 413

Tingley K. Lew

Damages Caused by Storm Surges in Maldives 431

Takaaki Uda

Performance of Structures in Hurricane Hugo* 434

Richard D. Marshall

The Peformance of Structures in Hurricane Hugo

1989 -The Carolinas* 445

Peter R. Sparks

Summary of Task Committee Workshop Reports 461

Task Committees "C" and "Z)" — Report of U.S. -Japan Workshop — Repair

Retrofit, and Evaluation of Structures Wind and Earthqake Resistance,

May 12-14, 1990 461

Task Committee "/" — Report on the Sixth U.S. -Japan Bridge Engineering

Workshop — Analysis, Design, Performance and Strengthening, Lake Tahoe,

Nevada, U.S.A., May 7-8, 1990 463

Appendix: Task Committees A-J Reports 467

* Papers presented during site visit at Charleston, South Carolina on Monday, May 21, 1989.

xii
AGENDA FOR 22nd JOINT UJNR MEETING

Tuesday 15 May

0930 Meet in Lobby of Holiday Inn for NIST Shuttle Bus

to Administration Building

1000 OPENING CEREMONIES (Lecture Room B, Administration

Building)

Call to order by Noel RAUFASTE, Secretary-General

US Side, UJNR Panel

Opening remarks by John LYONS, Director, National

Institute of Standards and Technology

Remarks by Takashi KISAKA, Counsellor for Science

and Technology, Embassy of JAPAN

Remarks by Richard N. WRIGHT, Chairman US Side,

UJNR Panel

Remarks by Toshio IWASAKI, Chairman JAPAN Side,

UJNR Panel

Introduction of US Members by US Panel Chairman

Introduction of JAPAN Members by JAPAN Panel

Chairman

Elect Joint Meeting Chairman

Adopt Agenda

Discuss Panel's contributions to IDNDR (Panel

Chairmen to review international statements

and respective JAPAN/US National Committee

activities)

Adjourn

1130 Group Photograph

1200 Lunch: Hosted by John LYONS, Director NIST

xiii
THEME - WIND ENGINEERING

1300-1500 Technical Session - Wind Engineering

Chairman: Toshio IWASAKI

1300-1320 Meteorological Data from Hurricane Hugo, Joseph H.

GOLDEN* , NOAA

1320-1340 Monitoring the Aerodynamic Performance of a

Classic Suspension Bridge, Harold R. BOSCH*, FHWA

1340-1400 Cooperative Research Program on Wind Tunnel

Testing in Turbulent Flow, Koichi YOKOYAMA*

,

Robert L. WARDLAW, Hiroshi SATO, Takaaki KUSAKABE,

Michael G. SAVAGE

1400-1420 Wind on Offshore Platforms, Charles SMITH*, MMS

1420-1440 Discussions

1440-1500 Break

(* identifies oral presenters)

TASK COMMITTEE MEETINGS

1500-1700 T/C Meetings

T/C "A", Employees Lounge

T/C "C" & "D", Lecture Room B

T/C "F", CBT Conference Room, Bldg 226, Rm B221

T/C "I", Building Materials Conf. Room, Bldg 226,

Rm. A3 68

1700 Conclusion of Day 1

1815 Meet at Lobby of Holiday Inn; travel by bus to

BLACKIE'S HOUSE OF BEEF, 22nd and M Street N.W.

,

Washington, DC

1930 US Members Hosted Dinner at BLACKIE'S HOUSE OF BEEF

2200 Return to Holiday Inn

xiv
Wednesday 16 May

0800 Meet in lobby of Holiday Inn for NIST Shuttle Bus

to Administration Building

THEME - STORM SURGE AND TSUNAMIS

0830-1010 Technical Session - Storm Surge and Tsunamis

Chairman: Richard WRIGHT

0830-0850 Deep Ocean Recordings of Tsunamis, Frank I. GONZALEZ*,

NOAA/PMEL

0850-0910 Measurement and Modeling of Coastal Storm Effects,

William E. ROPER*, H. Lee BUTLER, Andrew GARCIA, CORPS

0910-0930 Storm Surge Modeling in the United States, Chester

P. JELESNIANSKI*, Jye CHEN, Wilson A. SHAFFER, NOAA/NWS

0930-0950 Discussions

0950-1010 Break

THEME - JOINT COOPERATIVE RESEARCH

1010-1150 Technical Session - Joint Cooperative Research

Chairman: Richard WRIGHT

1010-103 0 Precast Concrete Structure, Shin OKAMOTO; Shinsuke

NAKATA*; Hisahiro HIRAISHI ; Takashi KAMINOSONO;

Hitoshi SHIOBARA; Mizuo INUKAI

1030-1050 U.S. Coordinated Program for Masonry Building

Research Fifth Year Status, James L. NOLAND (Presented

by H.S. LEW)

1050-1110 Design Guidelines of Medium Rise RM Buildings,

Yutaka YAMAZAKI; Shin OKAMOTO*; Tsuneo OKADA;

Akira MATSUMURA; Toshiyuki KUBOTA

1110-1130 U.S. -JAPAN Cooperative Research on Hybrid Control

of Seismic Response of Bridge Structures, Jiro

TAGUCHI; Toshio IWASAKI , Yoshio ADACHI ; Yasushi

SASAKI; Kazuhiko KAWASHIMA*

1130-1150 Discussions

1200-1300 Lunch: Hosted by Denis KWIATKOWSKI , Assistant

Associate Director, Office of Natural and Technological

Hazards Programs, FEMA

XV
THEME - EARTHQUAKE ENGINEERING

1300-1440 Technical Session - Earthquake Engineering-Part I

Chairman: Richard WRIGHT

1300-1320 Lateral Resistance of a Pile, Michio OKAHARA;

Shoji TAKAGI; Keiji TAGUCHI (Presented by Koichi

YOKOYAMA)

1320-1340 Major Seismic Building Upgrades at the Lawrence

Livermore National Laboratory, Frank J. TOKARZ

,

LLNL; Gary E. FREELAND* , LLNL; James R. HILL, DOE

1340-1400 Attenuation Characteristics of Ground Strains

Induced During Earthquakes, Ken-ichi TOKIDA;

Keiichi TAMURA* ; Koh AIZAWA

1400-1420 Discussions

1420-1440 Break

TASK COMMITTEE MEETINGS

1440-1700 Task Committee Meetings

T/C "B" Employees Lounge

T/C "E" Lecture Room B

T/C "H" Conference Room B113

T/C "J" Conference Room Bill

1700 Conclusion of day 2

1800 Individual US members hosted dinners

xvi
Thursday 17 May

THEME - EARTHQUAKE ENGINEERING

0830 Meet in Lobby of Holiday Inn for NIST Shuttle Bus

to Administration Building

0900-1200 Technical Session - Earthquake Engineering-Part II

Chairman: Toshio IWASAKI

0900-092 0 Dynamic Centrifugal Model Tests of Embankments on

Liquefiable Grounds, Yasuyuki KOGA; Junichi KOSEKI

(Presented by Keiichi TAMURA)

0920-0940 Seismic Resistance Requirements of Roller Compacted

Dams, Alan T. RICHARDSON* and Louis H. ROEHM,

BUREC

0940-1000 Damage of Soil Liquefaction Caused by Earthquakes

and Use of Geomorphological Maps, Takekazu AKAGIRI*

1000-1020 Discussions

1020-1040 Break

1040-1100 Dynamic Deformation Characteristics of Compacted

Rockfills by Cyclic Torsional Simple Shear Tests,

Norihisa MATSUMOTO; Nario YASUDA; Masahiko OKUBO;

Yasushi KINOSHITA (Presented by Kinj i HASEGAWA)

1100-1120 Allowable Residual Displacement of Gravity

Quaywall Given by Optimum Seismic Coefficient from

Economical Viewpoint, Tatsuo UWABE*

1120-1140 The Relation Between the Building Damages and the

Vibration Properties of Ground (in the case of

Spitak Earthquake), Keiichi OHTANI*; Hiroyoshi

KOBAYASHI

1140-1200 Discussions

1200-1300 Lunch: Hosted by Robert D. HANSON, Director,

Biological and Critical Systems Division, NSF

1300-1440 Technical Session - Earthguake Engineering Part III

Chairman: Toshio IWASAKI

1300-1320 New Seismic Design Specifications of Highway

Bridges in Japan, Toshio IWASAKI; Kazuhiko

KAWASHIMA; Kinj i HASEGAWA*

1320-1340 Scatter of Mechanical Properties of Structural

Steels Recently Supplied in Japan, Hirofumi AOKI

Hiroyuki YAMANOUCHI

1340-1400 Development on Effective Use of New Materials for

Building Structures, Hiroyuki YAMANOUCHI; Tatsuo

MUROTA (Presented by Hirofumi AOKI)

1400-1420 Discussions

1420-1440 Break

*

xvu
THEME - T/C WORKSHOP REPORTS

1440-1540 Technical Session - Summary T/C Workshop Reports

Chairman: Toshio IWASAKI

1440-1500 T/C "C&D" Workshop on Repair and Retrofit and

Performance of Structures (Presented by Shin OKAMOTO)

1500-1520 T/C "J" Workshop on Bridges (Presented by James

COOPER)

1520-1540 Discussions

1540 End of Technical Sessions

1540-1630 T/C Chairmen Meeting with Panel Secretaries to

review T/C meetings, T/C reports, and IDNDR.

1630 Conclusion of Day 3

1730 Depart Holiday Inn for Potomac River Cruise on the

Spirit of Washington , 6th and Water Streets S.W.,

Washington, DC

1830 Cruise on The Spirit of Washington

xviii
Friday 18 May

0800 Meet in Lobby of Holiday Inn for NIST Shuttle Bus

to Administration Building

THEME - LOMA PRIETA EARTHQUAKE

0830-1200 Technical Session - Loma Prieta Earthquake

Chairman: Richard N. WRIGHT

0830-0845 Initial Overview of the San Francisco Bay and

Santa Cruz Mountains Ground Motion, Gerald BRADY*,

USGS

0845-0900 Site Response and Liquefaction, Ellis KRINITZSKY

and Gus FRANKLIN*, WES/CORPS

0900-0915 Dam Performance During the Loma Prieta Earthquake

of October 17, 1989, Robert B. MacDONALD* , BUREC

0915-0930 Performance of Wood-Framed Structures in the Loma

Prieta Earthquake, John TISSELL*, American Plywood

Association

0930-0950 Discussions

0950-1010 Break

1010-1025 Seismic Performance of Nonstructural Elements

During the Loma Prieta Earthquake, T.T. SOONG*

,

NCEER and SUNY

1025-1040 Socioeconomic Impact of Lifeline Performance

Related to the Loma Prieta Earthquake, Ronald

EGUCHI* and M. PHIPPS, Dames and Moore

1040-1055 Lifelines Performance During the October 17, 1989

Loma Prieta Earthquake, M. KHATER and C. SCAWTHORN,

EQE Engineering, Jerry ISENBERG, Weidlinger

Associates, L. LUND, Consulting Engineer, T.

LARSEN, EQE Engineering, and Masanobu SHINOZUKA*,

Princeton University

1055-1115 Discussions

1115-114 5 Summary Report on the Loma Prieta Earthquake of

October 17, 1989, Toshio IWASAKI*, Kazuhiko KAWASHIMA

1145-1200 Discussions

1200-1300 Lunch: Hosted by Walter W. HAYS, Deputy Chief

Research Applications, Office of Earthquakes,

Volcanoes, and Engineering, USGS

XIX
TASK COMMITTEE REPORTS AND RESOLUTIONS

1300-1430 Task Committee Reports

Chairman: Richard N. WRIGHT

T/C A

T/C B

T/C C

T/C D

T/C E

T/C F

T/C H

T/C I

T/C J

Strong Motion Instrumentation Arrays and Data

Large-scale Testing Program

Repair and Retrofit of Existing Structures

Evaluation of Performance of Structures

Natural Hazard Assessment and Mitigation through

Land-use Programs

Disaster Prevention Methods for Lifeline Systems

Soil Behavior and Stability During Earthquakes

Storm Surge and Tsunamis

Wind and Earthquake Engineering for Transportation

Systems

1430-1445 Break

1445-1545 Adoption of Final Resolutions (including Panel

actions on IDNDR)

1545-1555 Break

CLOSING CEREMONIES

1555 Call to Order by N.J. RAUFASTE, Secretary- General

US Panel

Closing Remarks by Toshio IWASAKI, Chairman JAPAN

Panel

Closing Remarks by R.N. WRIGHT, Chairman US Panel

1615 Conclusion of Technical Sessions

1800 JAPAN-side hosted dinner at the Japan Inn, 1715

Wisconsin Avenue, N.W. Washington, D.C.

XX
LIST OF PANEL MEMBERS

UNITED STATES PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1990

Dr. Richard N. Wright

Chairman

Director, Center for Building Technology

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301.975.5900 FTS 879.5900 FAX 301.975.4032

Mr. Noel J. Raufaste

Secretary-General

Head, Cooperative Research Programs

Center for Building Technology

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301.975.5904 FTS 879.5904 FAX 301.975.4032

Dr. S. T. Algermissen

Research Geophysicist

Office of Earthquake Studies

Branch of Earthquake Tectonics

Geological Survey

U.S. Department of the Interior

Denver, CO 8 0225

303.236.1611 FTS 776.1519 FAX 303.236.1519

Dr. Celso S. Barrientos

Supervisory Physical Scientist

INESDIS-E/RA28

Suitland Professional Center, Room 315

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

5001 Silver Hill Road

Suitland, MD 20746

301.763.4178 FAX 301.420.1883

Dr. Eddie N. Bernard

Director, Pacific Marine Environmental Laboratory

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

7600 Sand Point Way, NE

BIN C15700/Building 3

Seattle, WA 98115-0070

206.526.6239 FTS 392.6800 FAX 206.526.6815

xxi
Dr. Roger D. Borcherdt

Branch of Engineering Seismology and Geology

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415.329.5619 FTS 459.5619 FAX 415.329.5163

Dr. A. Gerald Brady

Physical Scientist

Branch of Engineering Seismology and Geology

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415.329.5664 FTS 459.5664 FAX 415.329.5163

Mr. H. Lee Butler

Chief, Research Division

Coastal Engineering Research Center

U.S. Army Engineer Waterways Experiment Station

Office CEWES-CR

P.O. Box 631

Vicksburg, MS 39180-0631

601.634.2405 FTS 542.2405 FAX 601.634.4314

Dr. Ken P. Chong

Program Director of Structures and Building Systems

Directorate of Engineering

National Science Foundation

1800 G Street, N.W. , Room 1108

Washington, D.C. 20550

202.357.9542 FAX 202.357.7636

Mr. James D. Cooper

Deputy Chief, Structures Division, HNR-10

Federal Highway Administration

U.S. Department of Transportation

63 00 Georgetown Pike

McLean, VA 22101

703.285.2447 FAX 703.285.2379

Dr. A. G. Franklin

Chief, Earthquake Engineering & Geophysics Division

CEWES-GH Geotechnical Laboratory

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-3909

601.634.2658 FTS 542.2658 FAX 601.634.3139

xxii
Mr. G. Robert Fuller

Chief, Manufactured Housing Compliance Branch

Room 9156

U.S. Department of Housing and Urban Development

Washington, DC 20410-8000

202.708.1920 FAX 202.708.0299

Mr. James H. Gates

Caltrans - Structures

P. O. Box 942874

Sacramento, CA 94274

916.445.1439 FAX 916.445.0574

Mr. Peter E. Gurvin

Assistant Director for Buildings Design and

Engineering

Office of Foreign Buildings Operations

A/FBO, SA-6, Room 327

P.O. Box 12248

U.S. Department of State

Rosslyn Station

Arlington, VA 222 09

703.875.6117 FAX 703.875.6204

Dr. Robert D. Hanson

Division Director, Division of Biological &

Critical Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

202.357.9545 FAX 202.357.9803

Dr. Walter W. Hays

Deputy for Research Application

Office of Earthquake Survey

905 National Center

Geological Survey

U.S. Department of the Interior

Reston, VA 22092

703.648.6711 FAX 703 . 648 . 6717

Dr. William B. Joyner

Geophysicist

Branch of Ground Motion & Faulting

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415.329.5640 FTS 459.5640 FAX 415.329.5163

xxiii
Mr. Roger M. Kenneally

Structural Engineer

Structural and Seismic Engineering Branch

Seismic Section

Office of Nuclear Regulatory Research

U.S. Nuclear Regulatory Commission

Washington, DC 20555

301.492.3893 FAX 301.443.7836

Mr. James P. Knight

Office of Nuclear Safety (EH-33)

U. S. Department of Energy

Washington, D.C. 20545

301.353.3548 FTS 233.3548 FAX 301.353.5285

Mr. Richard W. Kramer

Technical Review Staff, D-3210

Bureau of Reclamation

U. S. Department of the Interior

P.O. Box 25007

Denver, CO 8 0225

303.236.8539 FTS 776.8539 FAX 202.647.5269

Mr. Paul F. Krumpe

Physical Scientist

Office of U.S. Foreign Disaster Assistance

Agency for International Development

U.S. Department of State

Washington, DC 20523

202.647.9758 FAX 202.647.5269

Dr . U.S. Lew

Chief, Structures Division

Center for Building Technology

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301.975.6061 FTS 879.6061 FAX 301.975.4032

Mr. Tingley K. Lew

Research Structural Engineer

Structures Division, Code L51

Naval Civil Engineering Laboratory

Port Hueneme, CA 9 3 043

805.982.1234 FAX 805.982.1418

Dr. Shih-Chi Liu

Program Director, Structural Systems

Division of Biological and Critical Systems

National Science Foundation

1800 G Street, N.W. , Room 1132

Washington, DC 20550

202.357.9780 FAX 202.357.9803

xxiv
Mr. Robert B. MacDonald

Manager, Geology Branch

Engineering and Research, Code D-3610 A

Bureau of Reclamation

U.S. Department of the Interior

Building 67, Denver Federal Center

Denver, CO 8 0225

303.236.6904 FTS 776.8643 FAX 303.236.6763

Dr. Richard D. Marshall

Structural Research Engineer

Structures Division

Center for Building Technology

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301.975.6071 FTS 879.6071 FAX 301.975.4032

Dr. Wayne N. Marchant

Chief, Research and Laboratory Services Division

P. O. Box 25007

Code D-3700

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 8 0225

303.236.5983 FTS 776.5983 FAX 303.236.7664

Dr. Francis G. McLean

Chief, Geotechnical Engineering and Embankment

Dams Branch

P.O. Box 25007

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303.236.3854 FTS 776.3854 FAX 303.236.6763

Mr. Herbert Meyers

Chief, Earth Geophysics

National Geophysical Data Center

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

325 Broadway

Boulder, CO 80303-3328

303.497.6521 FTS 320.6521 FAX 303.497.6513

Dr. William E. Roper

Assistant Director

Research and Development (Civil Works)

US Army Corps of Engineers

2 0 Massachusetts Avenue, N.W

Washington, DC 20314-1000

202.272.0257 FAX 202.272.0907

XXV
Dr. Erwin L. Schaffer

Assistant Director, Wood Products Research

Forest Products Laboratory

Forest Service

U.S. Department of Agriculture

One Gifford Pinchot Drive

Madison, WI 53705-2398

608.264.5672 FTS 364.5672 FAX 608.231.9592

Mr. Charles E. Smith

Research Program Manager

Offshore Minerals Management

Technology Assessment and Research Branch

Minerals Management Service

U.S. Department of the Interior

381 Elden Street

Herndon, VA 22070-4817

703.787.1559 FAX 703.787.1010

Dr. T. T. Soong

Samuel P. Capen Professor

Department of Civil Engineering

National Center for Earthquake Engineering Research

State University of New York at Buffalo

212 Ketter Hall

Buffalo, NY 14260

716.636.2469 FAX 716.636.3733

Mr. Kenneth F. Sullivan

Lifelines Project Manager

Earthquakes and Natural Hazards Division

Federal Emergency Management Agency

Washington, DC 20472

202.646.2889 FAX 202.646.3104

Mr. Edward F. Younger

Director, Structural Engineering Service

Code 088C1

Office of Facilities

Department of Veterans Affairs

810 Vermont Avenue, N.W.

Washington, DC 20420

202.233.2864 FTS 373.2864 FAX 202.233.7854

xxvi
ALTERNATE MEMBERS

Dr. Clifford J. Astill

Program Director

Division of Fundamental Research for Critical

Engineering Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

202.357.9500 FAX 202.357.9803

Dr. Mehmet K. Celebi

Research Civil Engineer

Branch of Engineering Seismology and Geology

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415.329.5623 FTS 459.5623 FAX 415.329.5163

Mr. Michael Changery

Chief, Applied Climatology Branch

National Climatic Center

National Oceanic and Atmospheric Administration

U. S. Department of Commerce

Federal Building

Asheville, NC 28801

704.259.0765 FTS 672.0765 FAX 704.259.0246

Mr. Vincent P. Chiarito

Research Structural Engineer

Structural Mechanics Division

Structures Laboratory

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39108-6199

601.634.2714 FAX 601.634.2309

Dr. James F. Costello

Senior Structural Engineer

Structural and Seismic Engineering Branch

Structural Section

Office of Nuclear Regulatory Research

U.S. Nuclear Regulatory Commission

Mail Stop 007NL

Washington, DC 20555

301.492.3818 FAX 301.443.7836

xxvii
Mr. Lucian G. Guthrie

Structural Engineer

Office of Chief of Engineers

HQUSACE (CEEC-ED)

U.S. Department of the Army

20 Massachusetts Avenue, N.W.

Washington, DC 20314-1000

202.272.8673 FAX 202.272.1485

Mr. James R. Hill

Manager, Natural Hazards Mitigation Programs

Office of Safety, Health and Quality Assurance

U.S. Department of Energy

Washington, DC 20545

301.353.4508 FAX 301.353.5285

Dr. James R. Houston

Chief, Coastal Engineering Research Center

US Army Engineer Waterways Experiment Station

P.O. Box 631

Vicksburg, MS 39180-0631

601.634.2000 FAX 601.634.2655

Mr. Robert R. Ledzian

Senior Staff Assistant for Research

Office of Liaison - Engineering Research, Code 3020

Bureau of Reclamation

U.S. Department of the Interior

18th & C Streets, N.W.

Washington, DC 20240

202.343.2013 FAX 343.2252

Mr. Ronald J. Morony

Director, Innovative Technology and Special

Projects Division

Office of Policy Development and Research

U.S. Department of Housing and Urban Development

Washington, DC 20410-6000

202.755.0640 FAX 202.755.0299

Dr. J. Eleonora Sabadell

Director, Natural and Manmade Hazards Mitigation

Program

Directorate of Engineering

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

202.357.9780 FAX 202.357.9803

xxviii
Dr. John B. Scalzi

Program Director for Structures and Building

Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

202.357.9542 FAX 202.357.7636

xxix
JAPAN PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1990

Dr. Toshio Iwasaki

Chairman

Director-General

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Yoshio Adachi

Secretary-General

Director, Planning and Research Administration

Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tadao Dohi

Head, Geographic Department

Geographical Survey Institute

Ministry of Construction

1, Kitazato, Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-64-1111

Mr. Tokunosuke Fujitani

Chief, The Second Research Laboratory

Applied Meteorology Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1 Nagamine, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-51-7111

Mr. Minoru Fuj iwara

Head, Bridge Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

XXX
Dr. Masaya Hirosawa

Deputy Director General

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Kazuhiko Kawashima * Head, Earthquake Engineering Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Yoshikazu Kitagawa

Director

Testing and Evaluation Department

Building Research Institute

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Mr. Yasuyuki Koga

Head, Soil Dynamics Division

Construction Method and Equipment Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel . 0298-64-2211

Dr. Norihisa Matsumoto

Head, Fill Type Dam Division

Dam Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tatsuo Murota

Director, Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151 -

xxxi
Dr. Shinsuke Nakata

Head, Structural Division

International Institute of Seismology and Earthquake

Engineering

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Setsuo Noda

Director, Structural Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka,

Kanagawa-ken 2 39

Tel. 0468-41-5410

Mr. Keiichi Ohtani

Head, Earthquake Disaster Prevention Laboratory

National Research Center for Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-51-1611

Mr. Shin Ohtsuka

Head, Typhoon Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1, Nagamine, Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-51-7111

Mr. Michio Okahara

Head, Foundation Engineering Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Shin Okamoto

Director, International Institute of Seismology and

Earthquake Engineering

Building Research Institute

Ministry of Construction

1, Tatehata, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

xxxii
Dr. Norio Oyagi

Director, Third Research Division

National Research Center for Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Tukuba-shi,

Ibaraki-ken 305

Tel. 0298-51-1611

Dr. Yasushi Sasaki

Director, Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Masaaki Seino

Head, Seismology and Volcanology Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1, Nagamine, Tsukuba-shi

Ibaraki-ken 3 05

Tel. 0298-51-7111

Mr. Yukitake Shioi

Director, Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Katsutoshi Tanimoto

Director, Hydraulic Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka,

Kanagawa-ken 2 39

Tel. 0468-41-5410

Mr. Ken-ichi Tokida

Head, Ground Vibration Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

xxxiii
Dr. Takaaki Uda

Head, Coastal Engineering Division

River Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Tatsuo Uwabe

Chief, Disaster Prevention Laboratory

Structural Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi

Kanagawa-ken 2 39

Tel. 0468-44-5030

Dr. Hiroyuki Yamanouchi

Head, Structural Dynamics Division

Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Yutaka Yamazaki

Head, Large-scale Structure Testing Division

Production Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Mr. Koichi Yokoyama * Head, Structure Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

* Secretary

xxxiv
LIST OF TASK COMMITTEE MEMBERS

Task CcHnmittee US Side

A. Stxong-Motion Instrumentation

Arrays and Data _

*A.G.

M.K.

A.F.

W.B.

R.B.

F.G.

H.

Brady

Celebi

Franklin

Joyner

MacDonald

McLean

Meyers

B. Large-scale Testing Programs *H.S.

C.E.

Lew

Smith

Repair and Retrofit of Existing

Structures

*K.P. Qiong

P.E. Gurvin

H.S. Lew

R.J. Morony

J.B. Scalzi

Evaluation of Structural

Performance

*G.R. Fuller

M.K. Celebi

R.D. Marshall

J.B. Scalzi

E. Natiiral Hazard Assessment and

Mitigation Through Land Use

Programs

*S.T. Algemissen

C . S . Barrientos

A.G. Brady

G.R. Fuller

R.D. Marshall

Japanese Side

*S. Noda

S. Hattori

TUT IchiJ^wa

Y. Ishiyama

K. untani

Y. Sasaki

TTn. Yamanouchi

*J\. Ohtani

wX . Koga

oo

.

Noda

o UKamouo

K. Yokoyama

H. Yamanouchi

''i'l. rlXI. vJodWo.

T"X . C Vp/Jveil L IV—./ ^V_/ H1 1 • kJl l-L-l lV_/i 1CU_ CI.

*S. Okamoto

T. Hanafusa**

K. Masainura**

T. Murota**

M. Okahara

H. Shinohara

H. Yamanouchi

K. Yokoyama**

*K. Kawashima

S. Hattori

T. Iwasaki

M. Nagaoka

Y. Sugimura

T. Uwabe

XXXV
Task CoimTdttee-continu^

F. Disaster Prevention Methods

for Lifeline Systems

H. Soil Behavior and Stability

During Earthquakes

I. Storm Surge and Tsunamis

J. Wind and Earthquake Engineering

for Transportation Systems

US Side Japanese Side

*S.C,

M.K.

J.D.

T.

J.B.

J.S.

K.F.

S.

Liu

Celebi

Cooper

Liu

Scalzi

Spencer

Sullivan

Wu

*Y. Sasaki

K. Kawashima

T. Murota

S. Noda

K. Ohtani

M. Shinozuka

T. Uwabe

*A.G. Franklin *K. Tc.LLda

E.G. McLean Y. Koga

C.E. Smith N. Matsumoto

M. Okahara

Y. Sugimura

H. Tsuchida

*H. Meyers *T. Uda

C.S. Barrientos M. Ichikawa

E.N. Bernard K. Kurashige

L. Butler K. Masaraura

M. Okada

N. Oyagi

H. TaJ^ahashi

K. Tanimoto

*J.D. Cooper *S. Saeki

A.G. Fr"anklin M. Fujiwara

J.H. Gates T. Iwasa3d.

J.B. Scalzi K. Kawashima

M. Okahara

Y. Sasaki

K. Yokoyama

*Chairman

February 1990

xxxvi
RESOLUTIONS

RESOLUTIONS OF THE TWENTY-SECOND JOINT MEETING

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS (UJNR)

National Institute of Standards and Technology

Gaithersburg, MD 20899

May 15-18, 1990

The following resolutions are hereby adopted:

1. The Twenty-Second Joint Panel Meeting provided an opportunity

to exchange valuable technical information which was

beneficial to both countries. In view of the importance of

cooperative programs on the subject of wind and seismic

effects, the continuation of Joint Panel Meetings is

considered essential.

2. The following activities have been conducted since the

Twenty-First Joint Meeting:

a. Guest researchers from both countries performed joint

research that advanced the state of wind and earthquake

engineering.

b. Technical documents, research reports and proceedings of

workshops were exchanged.

c. The Japanese MOC documents, on base isolation systems

for buildings were translated into English. They will

be reviewed by both sides for corrections and approval,

published, and disseminated to appropriate US and

Japanese engineering community members. The National

Technical Information Service of DoC will hold a press

briefing to promote the availability of these reports.

d. Workshops and a special Panel Theme were held:

a) Joint Workshop on Repair, Retrofit and Evaluation

of Structures, Task Committees (C) and (D) , at

Gaithersburg, MD, May 12-14, 1990.

b) Sixth Bridge Workshop, Task Committee (J) , at Lake

Tahoe, Nevada, May 7-12, 1990.

c) Held a special Panel Theme on "Loma Prieta

Earthquake of October 17, 1989" to summarize the

investigations performed by both sides.

3. Experiences from the post-Hurricane Hugo U.S. fact finding

team visits to the affected U.S. East Coast and to the Virgin

Islands on September 1989 were reported. Experiences from

the UJNR Coordinated Post-earthquake fact finding team's

xxxvii
visits to the sites of damaged structures from the October

17, 1989, Loma Prieta Earthquake were reported. The Panel

recognizes the importance of using these findings and

experiences and encourages similar exchanges of information

from future post-disaster investigations. Information

transfer will include reports,

seminars, and special Panel sessions as used at the TwentySecond Joint Meeting.

4. The Panel will continue to seek methods to contribute to the

International Decade of Natural Disaster Reduction (IDNDR)

such as exchanging Proceedings of Task Committees and Panel

Meetings with their respective Country's National Committees

of IDNDR, compiling bibliographies of research findings on

wind and seismic effects, and holding workshops and seminars.

This work will be performed by the U.S. and Japan Sides'

Secretaries

.

5. The Panel approves the redefinitions of the scope of T/C "E"

which redirects its emphasis to ground motion, microzonation,

and seismic design forces. The Panel also approves its name

change to Task Committee (E) on "Ground Motion and Seismic

Design Forces".

6. The Panel recognizes the importance of research being carried

out both in the U.S. and Japan on passive, active, and hybrid

control systems. At the recommendation of Task Committees

(B) , (F) and (J) , the Panel approves the establishment of a new Task Committee (T/C "G") on Passive, Active, and Hybrid

Control Systems.

7. The Panel recognizes the importance of the work by both sides

in the US-Japan Joint Research Program on Precast Seismic

Structural Systems (PRESSS) . Information about the program

will continue to be exchanged during the coming year.

8. The Panel accepts each Task Committee's report developed

during the Twenty-Second Joint Meeting. Each report presents

objectives, current and proposed scope of work, and past and

expected accompl ishments

.

9. The Panel endorses the following proposed Task Committee

workshops

:

a. Task Committee (A) is planning its 3rd Workshop on

Processing Strong Motion Records, with concentration on

array data. The workshop is targeted for May or June

1991.

xxxviii
b. Task Committee (C) is planning a workshop on the Design

Approaches and Construction Methods for Repair and

Retrofit of Buildings and Structures; May, 1991, in

Japan

.

c. Task Committee (D) proposes holding a workshop on

"Structural Performance Evaluation Methodologies" prior

to the 23rd Joint Panel Meeting in Tsukuba.

d. Task Committee (E) proposes holding a workshop on Ground

Motion and Seismic Design Forces in Japan in 1991.

e. Task Committee (F) proposes holding a workshop on

Disaster Prevention for Lifeline Systems, tentatively

scheduled in the U.S. in 1991.

f . Task Committee (H) proposes holding its second workshop

on Remedial Treatment of Potentially Liquefiable Soils,

in Japan, between October 1990 and March 1991.

g. Task Committee (I) will hold its workshop on Tsunamis,

in Honolulu in November 1990.

h. Task Committee (J) will hold its Seventh Bridge

Workshop, scheduled in Japan just prior to the 23rd UJNR

Joint Meeting.

Scheduling for these workshops shall be performed by the US

and Japan chairmen of the respective Task Committees with

concurrence of the Joint Panel chairmen. Results of each

activity shall be presented at the 2 3rd Joint Panel Meeting.

10. The Panel recognizes the importance of continued exchange of

personnel, technical information, research results, and

recorded data that lead to mitigating losses from earthquakes

and strong winds. The Panel also recognizes the importance

of using available large-scale testing facilities in both

countries. Thus, these activities should continue to be

strengthened and expanded and, as appropriate, share Task

Committee activities at other meetings that have technical

interests in the Task Committee activities. To facilitate

these exchanges, the Panel will provide official endorsement.

11. The Twenty-Third Joint Meeting of the UJNR Panel on Wind and

Seismic Effects will be held at PWRI , Tsukuba, Japan, May

1991. Specific dates, program, and itinerary will be

proposed by the Japan Panel with concurrence of the U.S.

Panel

.

xxxix

Theme I

Wind Engineering

Meteorological Data From Hurricane Hugo 1. INTRODUCTION

by

Joseph H. Golden'

ABSTRACT

We shall describe and examine critically

the meteorology of Hurricane HUGO as it

passed through the Eastern Caribbean on maximum surface winds and central pressure

place it as a Category 4 hurricane on the

Saf fir/Simpson scale. While it was approaching the Leeward Islands,

reconnaissance aircraft measured a recordtying 165 kts . Selected data from surface

sites, NOAA aircraft and satellites have

been gathered and will be used to

synthesize HUGO'S evolving dynamic

structure as it ravaged the U.S. Virgin

Islands and Eastern Puerto Rico. Peak

winds and rainfall amounts were within the

envelope of extrema produced by three

previous intense hurricanes which have

impacted the region this century. However,

we shall demonstrate that Hurricane HUGO

underwent some unusual mesoscale

oscillations in its track through the

Eastern Caribbean. Contrary to widespread

rumors in the news-media, we could find no

eviflence from an extensive aerial damage

survey of any tornadoes; however,

eyewitness accounts and damage assessments

both point to the periodic production of

destructive microbursts beneath the major

rainbands and eyewall of HUGO. Flooding

and landslides were minimal except for

serious urban flooding in San Juan, and

rainfall amounts up to 13-15 inches were

found in orographically favorable areas of

the islands. Finally, we shall summarize

overall forecasting performance for HUGO

in the Caribbean, and conclude that,

overall, official track forecasts and

warning lead-times were better than 10-year

averages for previous tropical cyclones

affecting this area. We conclude by

pointing out the degraded state of the

surface and upper-air observing network in

the region and suggest follow-up research

and needed improvements in the network for

both forecasters and researchers.

KEYWORDS: Meteorology; forecasting;

warning; hurricane; wind; pressure;

reconnaissance.

Office of the Chief Scientist, NOAA

Washington, D.C. 20235

Hurricane HUGO had its origins in a strong

tropical disturbance which moved off the

African coast on September 9, 1989. The

disturbance was accompanied by an area of

intense thunderstorms which was visible on satellite imagery. The official National

Hurricane Center (NHC) track for HUGO

begins on the lOth, when a tropical

depression formed to the southeast of the

Cape Verde Islands (see Fig.l) . HUGO moved

westward at a fast clip, 18 knots, across

the Eastern Atlantic, intensifying to

tropical storm stage on the 11th and full

hurricane on the 13th (Fig. 2 satellite

photo, when HUGO was about 1100 n mi east

of the Leeward Islands).

Hurricane HUGO therefore belongs to a class

of major hurricanes which has been termed

"Cape Verde" storms. These storms usually

originate from strong African wave

disturbances which intensify as they move

off the West African Coast and produce a tropical depression (in rare cases,

tropical storm) as they pass close to the

Cape Verde Islands. This early

intensification and vortex development

climatologically favors further

intensification, as the storm traverses a long stretch of very warm tropical Atlantic

waters at low latitudes (thus reducing

chances of weakening or early recurvature,

due to interactions with mid latitude

westerly flow aloft). Other notable Cape

Verde hurricanes of this century which have

affected Puerto Rico and the Virgin Islands

include the famous West Indian hurricane

of September 10-20, 1928; the "San Ciprian

storm" of 1932; the "Santa Clara storm" of

1956; Frederic and David, 1979; and Donna,

1960. The 1928 "San Felipe" hurricane

traversed the island of Puerto Rico from

the eastsoutheast and produce 30-year

record rainfalls (up to 25-30 inches), a lowest barometer reading of 27.50" at

Guayama , and recorded windspeeds gusting

to 160 mph (probable peak gust). Additional details on the San Felipe Hurricane

are given by Fassig (1928). That same

hurricane reintensif led after moving off

the northwest coast of Puerto Rico and

struck the Southeast Florida Coast as a category 4 storm on the Saf fir-Simpson

scale (Hebert et al , 1984). According to

Hebert and Taylor (1988), the September,

1928 hurricane was also the second

deadliest hurricane to affect the U.S.

during this century, killing 1836 people

in the Lake Okeechobee, Florida area (Table

1). Goodman (1989) did a statistical

analysis of these and other tropical storms

that affected Puerto Rico during 1886-1988 . He concluded (well before HUGO) that there

was at least a 50% probability of a tropical cyclone "hit" or "near miss" in

Puerto Rico during the 1989 season. Other

3
interesting meteorological aspects of

Hurricane HUGO, in the context of the 1989

Atlantic hurricane season as a whole, are discussed by Case (1990).

During the night of September 14, 1989,

Hurricane HUGO began to slow its forward

speed and turn more towards the westnorthwest. This was in response to a weakening

of the high pressure ridge to the north of

the storm. As sometimes noted by hurricane

forecasters when a hurricane slows its

forward motion in a favorable tropical

air/ocean environment, HUGO appeared to

strengthen appreciably on the 14th and

15th. A visible NOAA-GOES satellite photo

of HUGO on the 14th is shown in Fig. 3 (note

the well-developed eye).

2 . AIRCRAFT RECONNAISSANCE

When the first NOAA reconnaissance aircraft

was able to reach HUGO at mid-day on the

15th, the central pressure in the eye was

918 mb, and one-minute sustained windspeeds

of 165 kts were measured at 1500 ft

altitude. These data indicated that HUGO

had intensified to a rare category 5 status

on the Saf fir-Simpson scale (Hebert et al,

1984). Fig. 4 shows an enlarged visible

GOES satellite image of HUGO near the time

of the NOAA aircraft penetration. It

should be noted that the minimum pressure

found at this time ties the Atlantic record

(with Hurricane GLORIA, 1985, when it was north of Puerto Rico). Fig. 5a, 5b shows

the radar PPI reflectivity pattern measured

by the NOAA WP3D aircraft with its 5 cm

fuselage radar, and a NW-SE radar crosssection through the hurricane's eye

obtained with the 3 cm tail radar,

respectively (see Dodge et. al , 1990).

These remarkable photographs show that the

eyewall sloped outward slightly with

height, and had higher reflectivities

reaching greater heights (14 km) in the

northwestern sector than in the

southeastern portion of the wall-cloud.

Most unusual was the location of the

highest reflectivities (in excess of 50

dBz) in the southwest quadrant of the

eyewall. Close to the time of Fig. 5a, b,

one of the NOAA WP3D aircraft penetrated

the eyewall at 1500 ft (unfortunately, near

the reflectivity maximum in the

southwestern quadrant of the wall-cloud in

Fig. 5a). The aircraft-measured radial

plots of one-minute average tangential

winds and surface pressure are shown in

Fig. 6. Note that the pressure drops most

rapidly inward through the inner rainbands

and especially the wall-cloud of HUGO, to

a minimum of 918 mb in the eye itself.

Correspondingly, the extreme pressure

gradients in the same core region are

associated with windspeed maxima, at flight

altitude, of about 73 and 80 m/s (through

the SW and NE portions of the wallcloud,

respectively) . During this flight

penetration, the NOAA aircraft experienced

extreme turbulence and/or windshear, lost

one of its four engines (which caught

fire), and dropped from 1500 down to 800

ft before recovering flight control and

circling inside HUGO'S eye. Fortunately,

the aircraft was able to return safely to

land at the nearest airport on Barbados.

Subsequent inspection revealed no structural damage, and F. Marks and others

at NOAA/HRD/AOML in Miami are researching

the data taken during this and other

flights into HUGO; these data include

Doppler wind measurements made in the

rainbands of HUGO with the aforementioned

NOAA P-3 tail radar (see also Marks and

Houze, 1984, for Doppler analyses made on Hurricane Debby )

.

3. MESOSCALE VARIATIONS IN STORM

AND STRUCTURE

The well-developed eye of HUGO approached

the Lesser Antilles during the afternoon

of the 16th. A remarkable striated

structure in the cirrus outflow clouds

emanating from the hurricane's northern

portion is illustrated in the visible

satellite photo of Fig. 7. A time-series

plot of aircraft-measured surface central

pressures and eye diameters estimated from

aircraft radars is shown in Fig. 8. Note

that the time-history of surface pressure

in the eye begins near the time of the

record-equalling measurements on the

afternoon of the 15th—core pressures rose

appreciably during the following 24 hour

period. An unusual feature is that the

peak aircraft winds during the same period

remained nearly constant at about 140 mph.

Some of the problems in relating aircraft

reconnaissance winds, usually measured at

10,000 ft, to surface winds measured by

anemometers in hurricane have been

addressed by Powell (1980), Black et al

(1988) and Powell and Black (1990). Eye

diameters fluctuated during this same period until mid-day on the 16th, when Air

Force reconnaissance aircraft reported a double, concentric eyewall structure (the

significance of the double eyewall

structure for intense hurricanes was noted

by Willoughby et al, 1984 for a class of

hurricanes and Golden in the NAS/NRC Report

on Hurricane ALICIA, 1983 - see Savage et

al . 1984). The outer eyewall diameter of

HUGO was about 30 km, and the inner one was about 18 km across. An unusual feature of

the hurricane, as it passed through the

Lesser Antilles and Virgin Islands from

mid-day on the 16th through the morning of

the 19th (when it was northwest of Puerto

Rico) was the large fluctuations of the

eye's diameter (from 15 to 35 km across, see Fig . 8 )

.

4
4. IMPACTS ON THE ISLANDS

During the late night hours of Saturday,

September 16, the eye of Hurricane HUGO

passed over the island of Guadeloupe

(see Fig.l, near 16.4°N, 61.8°W). The

capital city, Pointe-A-Pitre , reported a final weather observation of heavy showers

and thunderstorms with NNW winds of 29 kts,

gusting to 84 kts. At the same time the

island of St. Maarten reported NE winds ISIS kts, and St. Kitts, north winds 21 kts

gusting to 33 kts. Reports after HUGO'S

passage over Guadeloupe revealed that about

half the capital was destroyed, including

the airport and there were 11 fatalities

(Table 2). HUGO continued moving westnorthwestward into the Caribbean Sea

as a category 4 hurricane with peak sustained winds of 140 mph (as measured by reconnaissance aircraft). Besides

Guadeloupe, the smaller island of

Montserrat to the northwest suffered severe

damages and 10 people were killed.

The storm slowed its forward motion to 8

kts during the day of the 17th and deepened

about 15 mb (flight-level maximum winds

from reconnaissance aircraft fluctuated

from 95-145 kts during the same period).

By the evening of the 17th, both satellite

and aircraft eye fixes indicated that HUGO

was turning more towards the northwest,

thus increasing the threat to the U.S. and

British Virgin Islands, especially St.

Croix and St. Thomas, as well as the

northeast coast of Puerto Rico. A basemap

of key islands and city locations cited in

the text is given in Fig. 9. A GOES

satellite view of HUGO on the afternoon of

Sept. 17 is given in Fig.lO.

St. Croix clearly experienced an unusually

prolonged battering of hurricane force (and

much greater) winds from late evening on the 17th into the next morning. It was noted in the 9; 00 p.m. AST advisory from

the National Hurricane Center that St.

Croix reported sustained winds of 85 mph

and gusts to 97 mph, with St. Thomas

reporting gusts to 90 mph during the

preceding hour. Both islands further

reported wind gusts of lOO mph during the

following hour. Fig. 11 shows a mesoscale analysis of HUGO'S track through the U.S.

Virgin Islands and Eastern Puerto Rico, derived from San Juan's NWS radar film and

NHC's official fixes. Note that by early

afternoon on the I7th, in response to

subtle, poorly-observed large scale changes

in HUGO'S environment (noted above), the

storm took a turn toward a more

northwesterly course at 12 mph. The

impacts of the hurricane's environment on

Its present and future motion and the density of observations on that scale were

considered by Chan et al (1980) and Peak

et al (1986). The eye slowed and executed

a trochoidal loop near Frederiksted , St.

Croix at around 0200 AST Sunday, September

18 (Fig. 11), before slowly curving

westnorthwestward (also noted by Lawrence

and Mayfield, 1977 for another hurricane).

The eye also underwent some interesting

structural changes as it approached St.

Croix. These are documented in Fig. 12, a sequence of photos from the NWS 10 cm radar

at San Juan. Note the contractions and

expansions of the eye and the position of

the northern eyewall over St. Croix for

several hours late on the 17th into the

18th. Storm, damaging winds resumed over

St. Croix after eye passage around 0400 AST

on September 18. A composite analysis

derived from airborne 5 cm radar data and

flight-level (700 mb) winds on the NOAA P3 aircraft, as HUGO was approaching iSt

.

Croix on the afternoon of the 17th, is

shown in Fig. 13. a. Note the well-defined

circular wall cloud with strongest

reflectivities in the NW and SE quadrants,

and a swath of 135 kt maximum windspeeds

around the NE and E quadrants in Fig. 13.

All surface observations from St. Croix and

St. Thomas airports ceased before midafternoon on Sunday due to extensive wind

damage and communication outages. Our

overall assessment of structural damages

and probable surface windspeeds is given

by Marshall (1990) in a companion paper.

We can now understand the extensive nature

and severity of the damage on St. Croix

from examination of Figs. 11 and 12.

After pummeling the U.S. Virgin Islands for

several hours, Hurricane HUGO slowly looped

into Vieques Sound during the predawn hours

of September 18, between the islands of

Culebra and Vieques (Figs. 9, 11). Note

especially the 2 well-defined trochoidal

loops in HUGO'S track on Fig. 11. Similar

hurricane track motions were documented for

hurricane ALICIA, 1983, prior to its

destructive landfall on Galveston Island,

Texas (Savage et. al. 1984) and for

Hurricane Carla, 1961. The most severe

surface damage found by our study team from

HUGO was on the islands of Culebra and St.

Croix, both of which were just to the right

of the cusp-points in Hugo's track (see

Fig. 11 ) . The damage on Vieques, to the left of the

track, was not quite as extensive in scale

but was almost as severe. After brushing

the eastern portion of Vieques at around

0730 AST on the 18th, HUGO'S eye looped

slowly westward over the northeastern

portion of Puerto Rico between 0800 and

0900 AST. Satellite data and San Juan

radar (Fig. 12) indicate that the eyewall

on the west side of the eye moved over land

near the towns of Ceiba (Roosevelt Roads

NAS ) , Fa jardo and Luquillo ( see Figs . 9 , 11 ) , while the east side of the eye remained

over water. It was during the last few

hours of HUGO'S approach to Puerto Rico

that Culebra probably experienced the

5
worst wind effects, with the southeasterly

flow (associated with the most dangerous

northeast quadrant of HUGO'S eyewall,

Fig. 12) channeled through the hills on both

sides of the Ensenada Honda Harbor (Fig. 9) . HUGO'S ill-defined eye appeared to bounce

NNW off the northeast coast of Puerto Rico

(Fig. 11) and by noon on Monday, September

18, the eye was over water north of San

Juan with maximum sustained winds of 125

mph and minimum sea-level pressure of 957

mb. The storm's radar structure, as documented by the 5 cm belly radar from the

NOAA WP3D aircraft is shown in Fig. 13. b.

Note that the southeastern half of the

storm is nearly devoid of rainbandsl

An analysis of the storm-total

precipitation from HUGO over Puerto Rico

is shown in Fig. 14. Note that maximum

rainfall amounts of 10-14 inches are found

in the mountainous terrain (3,000 ft

maximum heights), near the center of the

island. These maxima occurred in

orographically-favorable locations (compare

with Figs, 11 and 12) which experienced

primarily strong upslope flow during HUGO'S

passage. These rainfall amounts agree will

with those forecasted in the official NHC

advisories and bulletins on Hurricane HUGO

as it was approaching.

5. SOME UNIOUE DATA

A storm chaser, Jim Leonard, was able to

position himself in a multi-story

condominium in Luquillo, and produced a remarkable videotape of the approach and

passage of HUGO'S eye directly over him.

The videotape documents damaging wind and

rain effects on nearby structures during

major rainband and eyewall passages in

HUGO, as well as chaotic state of the

adjacent sea surface. He used a digital

barometer to measure a lowest pressure. San

Juan, which remained outside the eye,

recorded a minimum pressure of 970.3 mb.

The radar sequence in Fig. 11 supports the

pressure data indications that HUGO was

filling as it crossed the northeast coast

of Puerto Rico. However, we must emphasize

that the western eyewall passed just to the

east of the metropolitan San Juan, probably

affecting Loisa and Pinones (see Figs.

9

and 11); moreover, this geometry is

entirely consistent with the large

gradations of damage and surge effects

(especially overwash) documented by the

team from Catano eastward. Our assessment

of the damage to homes and commercial

buildings and probable associated windspeed

maxima is given in Marshall (1990).

A very unique aspect of Hurricane HUGO

occurred on the afternoon of the 18th, when

the storm was north of San Juan (Fig. 13).

During our interview with the

meteorologist-in-charge of the San Juan

WSFO, Israel Matos , he made the point of

recalling that he and some of his staff had

felt an earthquake on Puerto Rico on Monday

afternoon. We have since confirmed that

a magnitude 4.8 (Richter scale) earthquake

occurred at 1447 AST and it was centered

in the U.S. Virgin Islands. This was certainly a significant earthquake event

in a seismically-active area, and perhaps

further research is warranted on the

possible linkages to HUGO, if any.

6. OBSERVATIONAL GAPS

The team found an appalling lack of

recorded meteorological surface data,

especially winds, over Puerto Rico and the

U.S. Virgin Islands in HUGO'S path. As

noted in other recent NAS/NRC hurricane

study-team reports (e.g. Savage et_a_l, 1984 on Hurricane ALICIA) , many anemometers were

in place, but most were either damaged or destroyed by HUGO'S winds, or lacked

recording equipment and/or back-up power.

Only two wind records have been obtained

so far: one from the National Weather

Service (NWS) Forecast office at San Juan

Munoz Airport and the other at Roosevelt

Roads NAS both sites on Puerto Rico

(Figs. 9, 14). A time history of peak wind

gusts for these sites is shown in Fig. 15

a,b, resp. Recall that San Juan Airport

probably experienced the fringes of HUGO'S

western eyewall between (1300-1400 GMT,

while Roosevelt Roads NAS experienced the

eye itself about 1 1/2-2 hours earlier).

The NWS Forecast Office at San Juan

recorded peak gusts of 92 mph between 1350

and 1415 GMT, and the maximum sustained

wind speed was 77 mph. Adjustment for the

6.1 meter height of the F420 C anemometer

would increase the plotted windspeed in

Fig. 15a by about 7 percent. The fastest

mile speeds for standard conditions were

derived from the measured 10-minute mean

speeds using the procedure described by

Marshall (1984), and the maximum value of

88 mph occurred at 1350 GMT. Similarly,

the anemometer site at Roosevelt Roads is

well-exposed, and height of the

propeller/vane anemometer is 7 meters. The

time history of recorded peak gusts,

unadjusted for anemometer height in Fig.

15b shows peak gusts of 120 mph occurred

between 1150 and 1220 GMT. This site

experienced a brief passage of the

hurricane's eye at 1250 GMT (dip in

windspeeds) . Maximum sustained windspeeds

were about 98 mph. Fastest-mile speeds

were derived using the procedure used for

the San Juan Airport (NWS) data, and for

Roosevelt Roads the maximum value of 110

mph occurred at 13 20 GMT.

We have had to reconstruct probable oneminute sustained windspeeds and peak gusts

for the other nearby islands, using our

aerial damage survey and reduction of

reconnaissance aircraft-measured winds at

6
10,000 ft. down to the surface, using new research results of Powell and Black

(1989). These surface windspeed estimates

for Hurricane HUGO are summarized in

tabular form by Marshall (1990) . There was

an unofficial estimate of winds gusting to

150 kts in the harbor at Culebra (by a mariner, Mr. Herbert, who rode out the

storm in his sailboat and videotaped his

anemometer). Therefore, overall, we believe that the highest winds occurred on Culebra and St. Croix (estimated peak gusts

to 67 and 70 m/s , respectively. Another

unique data set has recently come to light:

a mini-network of six anemometers installed

by the FAA at San Juan Airport survived

HUGO'S passage. This system, called the

Low-level Wind Shear Alert (LLWAS) system,

has been installed at a number of airports

across the U.S. and its territories to warn

pilots of dangerous wind shear and

downburst events in the airport terminal

area. An LLWAS anemometer system was in

place at Houston's Hobby Airport during the

passage of Hurricane ALICIA, 1983 but the

NAS/NRC team report noted the failure of

the system to provide any wind data during

ALICIA due to a lack of recording equipment

and back-up power by FAA program managers.

Fortunately, the team's recommendations

were heeded, and we understand that wind

data were recorded during most of HUGO,

from 4 of 6 LLWAS sensors on San Juan

Airport. One of the sensors measured a peak gust of 43 kts. These data, combined

with a more complete record from NWS San

Juan airport, should provide a unique

microscale (few km domain) analysis of

surface wind distribution near the western

eye-wall passage of hurricane HUGO during

its temporary filling stage.

FORECAST PERFORMANCE

The team found that, overall, the accuracy

and timeliness of most of the official

bulletins and advisories issued by NHC and

Hurricane Local Statements by NWS San Juan

during Hurricane HUGO'S approach and

passage through the Eastern Caribbean were

state-of-the-art. This is especially so considering the paucity of hourly surface

and twice-daily upper air reporting

stations in the area (there are fewer

regularly-reporting upper sites on the

Eastern Caribbean islands today than 20

years ago) . The warning lead times for

the various islands are shown in Fig. 16,

and clearly indicate that most places had

at least 1-2 days warning of HUGO'S

approach. These lead times were

significantly better than is often possible

for hurricanes approaching landfall along

the U.S. mainland coast.

The hurricane forecasters at NOAA's

National Hurricane Center in Miami produce

an official forecast track, updated at sixhourly invervals, throughout the

hurricane's life-cycle. Fig. 17 shows the

final smoothed verifying track for

Hurricane Hugo (derived from all NOAA and

U.S. Air Force reconnaissance aircraft

fixes and satellite eye positions-bold

line) and the official forecasted tracks.

The official forecast track is produced by

the hurricane forecaster after he

subjectively assesses the outputs from a dozen or so numerical models and guidance

products. Ward (1990) has assessed the

performance of the models run on HUGO by

the NWS' National Meteorological Center.

Our Table 3 shows the mean errors for 6 of

the models routinely used by NHC—with only

a few exceptions the recently-revised

statistical/dynamical model (NHC83)

developed by Neumann and his colleagues at

NHC consistently outperformed all the other

models, even the more sophisticated

dynamical ones such as SANBAR (barotropic) , BAM (beta-advection model), and the QLM

( quasi-Lagrangian model). The interested

reader is referred to Ward's (1990) article

for more details on the models. This

outcome may, we believe, have more to do

with the state of the observational

database in the Caribbean and techniques

used to initialize the models than with any

inherent scientific deficiencies in the

models themselves.

Note that as HUGO approached the Leeward

Islands and entered the Eastern Caribbean,

there was a persistent left bias in the

official NHC forecasts for the hurricane's

future track in that region. This led to

initial forecasts for HUGO'S landfall on the south coast of Puerto Rico which were

not corrected until Sunday night, September

17. Nevertheless, both Mr. Matos , the MIC

at the San Juan NWS forecast Office and the

forecasters at NHC were aware of this bias,

so that the people living in the U.S.

Virgin Islands had plenty of time to make

all necessary preparations; most

importantly, there was also an early

decision, based on updated NHC advisories

with rising hurricane strike probabilities,

to evacuate people in San Juan to shelters.

We must emphasize that current hurricane

track forecasting models and techniques

available to NHC cannot forecast mesoscale

changes or oscillations in hurricane tracks

as depicted for HUGO in Fig. 11 earlier.

Moreover, there are currently no forecast

models for predicting the intensity changes

for HUGO implied by the eye-pressure

changes shown in Fig. 6.

We have found several locations in our

aerial damage survey over the U.S. Virgin

Islands where the debris patterns strongly

suggest the production of destructive

"microbursts" (Fujita, 1985). Fig. 18a,

b

show good examples of mircoburst damage,

as seen from our low-flying aircraft, over

St. Croix and Vieques. The most

distinctive features of raicroburst events

7
are the narrow confines and divergent

nature of the debris trails from destroyed

structures. Fujita (1978) found that much

of the severe damage to structures he found

in Corpus Christi , Texas from Hurricane

CELIA, 1971 was due to the formation of

intense microbursts (surprisingly), in the

western eyewall and a few major rainbands,

as the storm made landfall. We have found

no evidence of any tornadoes from either

the aerial or the ground damage surveys we performed on Puerto Rico and the U.S.

Virgin Islands. In any case, it is likely

that the general wind damage caused by the

hurricane itself, and those more concentrated damage areas from microbursts, both

would tend to mask any separate tornado

damage tracks from subsequent identification.

7. CONCLUSIONS AND RECOMMENDATIONS

One of the major issues in HUGO'S

aftermath in the Caribbean has been the

appalling lack of recorded surface

windspeeds and other meteorological data

(especially pressure and rainfall).

Without these critical measurements in the

future, there can be no resolution to the

question of correlation of the observed

damage of structures with the actual wind

loads generated by hurricanes. In

addition, there is a continuing need for

operational and research aircraft

reconnaissance into hurricanes, but there

is also an urgent need for more rugged,

reliable surface instrumentation to

properly calibrate and reduce the aircraft

data gathered on hurricanes approaching

landfall to appropriate surface values.

Moreover, surface recording stations are

too sparse in the Caribbean to measure the

actual surface wind maxima at the point

of hurricane landfall. This situation is

compounded by the fact that some of the

most important surface observing sites

have been closed by the Caribbean nations

involved because of high manpower and

maintenance costs.

RECOMMENDATION 1. Develop and deploy high

quality, automated surface observing

instruments which are rugged enough to

withstand hurricane wind speeds, and of

affordable cost. The Australians have

developed a relatively inexpensive device

for measuring maximum windspeeds (and wind

directions), called the "Maxometer", which

they have deployed along sections of their

hurricane-prone coastlines. We further

recommend that responsibility to deploy

and maintain instruments during local

hurricanes be given to selected

universities in the coastal States and on islands in the Caribbean with engineering

or meteorology programs.

We have also found that, even though the

official forecasts and warnings for HUGO

in the Caribbean were entirely adequate

and state-of-art , there is a need for

improvements in our hurricane forecasting

capability. However, such efforts will

have limited potential for breakthroughs

because there is a lack of upper-level

wind observations of the hurricane's

environment. Complete vertical wind

profiles are available only twice per day

from a shrinking number of upper air

rawindsonde sites in the Eastern

Caribbean. These are supplemented by cloud

motion derived estimates of layer-mean

winds aloft from satellite cloud tracking

(only from cloudy to partly cloudy

regions )

.

RECOMMENDATION 2. We need to install more

conventional rawinsonde stations, perhaps

in a cost-sharing arrangement with the

Caribbean countries involved. A better

long-term solution, and one with minimal

manpower requirements, will be to install

wind PROFILERS on selected islands. These

instruments are rugged, operate in all

weather continuously, and have been

installed on a few South Pacific island

sites as part of the TOGA climate program.

Finally, there needs to be research to

better define and exploit the NEXRAD

Doppler radar capabilities for estimating

surface windspeeds in hurricanes (we note

that a NEXRAD installation is planned for

San Juan , PR)

.

RECOMMENDATION 3 . There needs to be a systematic approach and sustained, longterm effort at developing improved

hurricane prediction models which give

consistent results for future track and/or

intensity changes in different regions.

The current situation is one where the

hurricane forecaster must subjectively

pick the hurricane model of the day, based

on recent performance for the particular

hurricane in a particular area (there are now at least 10-12 different hurricane

prediction models and techniques in

operation or quasi-operational (test) use

by NHC and the National Meteorological

Center )

.

Finally, we have found that lifelines were

particularly susceptible to total loss or prolonged disruption from HUGO'S course

through Puerto Rico and the Virgin

Islands. Power and telephone lines were

downed everywhere, mainly from toppled

trees and flying debris. Future

mitigation efforts should stress the

importance of putting utilities

underground whenever possible (already

done on the island of Tortola). There

also needs to be a careful reexamination,

after HUGO, of the probability of

windspeeds in excess of current codes in

the Caribbean to determine if code

revisions are needed.

8
There continues to be great pressure for

shoreline development (commercial as well

as residential) on hurricane and other

storm prove shorelines in Puerto Rico and

the Virgin Islands. There will be a continuous threat of beach erosion and

overwash, and other storm surge and wave

impacts on these exposed coastal

properties. Indeed, our team found that

HUGO'S surge and wave impacts on the

developed coastal areas of northeastern

Puerto Rico have made that region even

more susceptible to future damage from

winter storms in the Atlantic.

RECOMMENDATION 4 . Fund impact studies on offshore mining of sand for beach

replacement and determine the cost-tobenefit ratios. In addition, fund

programs to evaluate and improve, when

necessary, codes and enforcement for

beach-use management. Finally, support

and fund programs for unified, long-range

planning for public/private/government use

of existing beach resources.

8. ACKNOWLEDGEMENTS

All of the material in this paper is

derived from my work as Team Leader of a guick-response study team organized by the

National Research Council's Committee on Natural Disasters and dispatched to Puerto

Rico and the Virgin Islands within 3 days

of HUGO'S onslaught there. The other team

members who have contributed greatly to the

overall effort and much of this paper are:

R.D. Marshall, NIST; J. L. Vogel , NWS/NOAA,

Ben Aguirre, Texas A&M University, and

David Bush, Duke University. I am grateful

to my colleagues Peter Black, Frank Marks,

Mark Powell and Hugh Willoughby,

HRD/AOML/NOAA for sharing some of their

excellent data sets and preliminary

analyses with me; likewise, additional

lagniappes go to Miles Lawerence and Bob

Case of NHC/NWS/NOAA for many of the

operational data sets on HUGO, and to Jack

Parrish, AOC/NOAA for some of the NOAA P3 figures. Satellite data were kindly

provided on loan by Dane Clark,

NESDIS/NOAA. The manuscript was ably typed

by Donna Robertson, NOAA/CS.

9. REFERENCES

1. Baynton, H.W. (1978). "Radar Design

for Determining the Strength of

Tropical Cyclones in the Bay of Bengal

1976." Monthly Weather Review, 105

f 11) . 1458-1461.

2. Black, P.G., R. L. Elsberry, L. K.

Shay, R. P. Partridge, and J. D.

Hawkins (1988). "Atmospheric Boundary

Layer and Oceanic Mixed Layer

Observations in Hurricane Josephine

Obtained from Air-Deployed Drifting

Buoys and Research Aircraft." J.

Atmos. Oceanic Tech., 5(6) . 683-695.

3. Carbone, R. E. and F. D. Marks, Jr.

(1989). "Velocity Track Display

(VTD): A Real-Time Application for

Airborne Doppler Radar Data

Hurricane." AMS Conference on Hurricanes and Tropical Meteorology,

San Diego, CA, pp. 11-12.

4. Case, R (1990). "Hurricane-Strong

Storms Out of Africa." Weatherwise . 41(1) February, 1990, 23-29.

5. Chan, J. C. L. , W. M. Gray, and S. Q.

Kidder (1980). "Forecasting Tropical

Cyclone Turning Motion from

Surrounding Wind and Temperature

Fields." Mon, Wea . Rev. , 108 . 778-

792.

6. Dodge, P. P., J. S. Griffin, F. D.

Marks, Jr., and R. W. Burpee (1990).

"Interactive Analysis of NOAA P-3

Aircraft Data Support of Operational

Hurricane Forecasting." Preprints , Sixth International Conf . on Interactive Info, and Proc . Systems

for Meteor., Oceano. , and Hydrology,

Amer. Meteor. Soc, Boston, MA, 3 pp.

7. Fassig (1928). "San Felipe-the

Hurricane of September 13, 1928, at

San Juan, P.R. " Mon. Wea. Rev. . 56(9) . 350-352.

8. Fujita, T.T. (1978). "Manual of

Downburst Identification for Project

NIMROD." SMRP Research Paper 156.

Dept. of Geophys. Sci., U. of Chicago,

104 pp.

9. , (1985). The Downburst,

Microburst, and Macroburst . The Univ.

of Chicago,

122 pp.

10. Goodman, R.R. (1989). "An Historical

Look at Tropical Cyclone Paths in

Relation to Puerto Rico and the Virgin

Islands." Technical Note , National

Weather Service Southern Region

Headquarters, Ft. Worth, TX , 6 pp.

11. Hebert, P. J., and G. Taylor (1988).

"The Deadliest, Costliest, Most

Intense U.S. Hurricanes of this

Century (and other Frequently

Requested Hurricane Facts)." NOAA

Technical Memorandum NWS NHC 18 . Nat'l

Hurricane Center, Miami, FL, 28 pp.

12. , , and R.A. Case

(1984) . "Hurricane Experience Levels

of Coastal County Populations - Texas

to Maine." NOAA Tech. Memo. NWS NHC

24, Nat'l Hurricane Center, Miami, FL,

127 pp.

13. Jarvinen, B. R. , and M. B. Lawrence

(1985) . "An Evaluation of the SLOSH

Storm-Surge Model . " Bull. Amer.

Meteor. Soc. . 66( 11) . 1408-1411.

14. Lawrence, M. B. , and B. M. Mayfield

(1977)." Satellite Observations of

Trochoidal Motion During Hurricane

Belle 1976." Monthly Weather Review.

105(11) . 1458-1461.

9
15. Marks, F. D. , Jr. and R. A. Houze,

Jr. (1984). "Airborne Doppler Radar

Observations in Hurricane Debby."

Bull. Amer. Meteor. Soc. . 65(6) , 569-

582.

16. Marshall, R.D. (1984). "Fastest-Mile

Wind Speeds in Hurricane Alicia. "NBS

Technical Note . No. 1197, NBS,

Gaithersburg , MD

17. , R.D. (1990). "Performance

of Structures in Hurricane HUGO."

Paper presented at 22nd Joint Meeting,

UJNR, Panel on Wind and Seismic

Effects, Gaithersburg, MD.

18. Peak, J. E. , W. E. Wilson, R. L.

Elsberry, and J. C.-L. Chan (1986).

"Forecasting Tropical Cyclone Motion

Using Empirical Orthogonal Function

Representations of the Environmental

Wind Fields." Mon. Wea. Rev. , 114 , 2466-

19. Powell, M.D. (1980). "Evaluations of

Diagnostic Marine Boundary-Layer

Models Applied to Hurricanes." Mon.

Wea. Rev. . 108 . 757-765.

20. , and P. G. Black (1990).

"The Relationship of Hurricane

Reconnaissance Flight-Level Wind

Measurements to Winds Measured by

NOAA's Oceanic Platforms." Submitted

to J. Wind Enqr. , 12 pp.

21. Ward, J. H. (1990). "A Review of NMC

Numerical Forecast Guidance for

Hurricane HUGO. " Submitted to Weather

and Forecasting , Amer. Meteor. Soc,

Boston, MA.

22. Willoughby, H. E., F. D. Marks, Jr.,

and R. J. Feinberg (1984).

"Stationary and Moving Convective

Bands in Hurricanes." J . Atmos . Sci . . 41(232) . 3189-3226.

10
Table l . The most intense United States hurricanes of this century (at

time of landfall), ^ft^r Hebert and Taylor(1988),

MOST INTENSE HURRICANES, UNITED STATES 1900-1982

(At time of landfall)

VP A I?

1. Florida (Keys) 1935 5 892 26.35

2. CAMILLE (La. /Miss.) 1969 5 909 26.^4

3. Florida (Keys) /South Tex. 1919 4 927 27.37

4. Florida (Lake Okeechobee) 1928 4 929 27.43

5. DONNA (Fla. /Eastern U.S.) 1960 4 930 27.46

6. Texas (Galveston) 1900 4 931 27.49

7. Louisiana (Grand Isle) 1909 4 931 27.49

8. Louisiana (New Orleans) 1915 4 931 27.49

9. CARLA (Texas) 1961 4 931 27.49

10. Florida (Miami) 1926 4 935 27.61

11. HAZEL (S.C./N.C.) 1954 4* 938 27.70

12. Southeast Fla. /La. -Miss. 1947 4 940 27.76

13. North Texas 1932 4 941 27.79

14. AUDREY (La. /Tex.) 1957 4# 945 27.91

15. Texas (Galveston) 1915 4# 945 27.91

16. CELIA (S. Texas) 1970 3 945 27.91

17. ALLEN (S. Texas) 1980 3@ 945 27.91

18. New England 1938 : 3* 946 27.94

19. FREDERIC (Ala. /Miss.) 1979 3 946 27.94

20. Northeast U.S. 1944 3* 947 • 27.97

21. S. Carolina/N. Carolina 1906 3 947 27.97

22. BETSY (Fla. /La.) 1965 3 948 27.99

23. Southeast and Northwest Fla. 1929 3 948 27.99

24. Southeast Florida 1933 3 948 27.99

25. South Texas 1916 3 948 27.99

26. Miss. /Ala. 1916 3 948 27.99

27. South Texas 1933 3 949 28.02

28. BEULAH (S. Texas) 1967 3 950 28.05

29. HILDA (Louisiana) 1964 3 950 28.05

30. GRACIE (S. Carolina) 1959 3 950 28.05

31. Texas (Central) 1942 3 950 28.05

32. Southeast Florida 1945 3 951 28.08

33. Florida (Tampa Bay) 1921 3 952 28.11

34. CARMEN (Louisiana) 1974 3 952 28.11

35. EDNA (New England) 1954 3* 954 28.17

36. Southeast Florida 1949 3 954 28. 17 37. ELOISE (Northwest Fla.) 1975 3 955 28.20

38. KING (Southeast Fla.) 1950 3 955 28.20

* Moving more than 30 miles per hour.

# Classified category 4 because of extreme tides.

(3 Reached Cat. 5 intensity three times along its path through the Caribbean

and Gulf of Mexico. The lowest pressure reported was 899 mb (26.55 in.)

at 1742Z 8-7-80 off the northeastern tip of Yucatan Peninsula.

Table 2 . Estimated number of deaths associated with Hugo

.

South Carolina 13 Antigua and Barbuda 1 North Carolina 1 Guadeloupe 11

Virginia 6 Montserrat 10

New York 1 St. Kitts and Nevis _A

Puerto Rico 2 TOTAL 49

U.S. Virgin Islands 3

11
Table 3.

HURRICANE HUGO

TRACK ERRORS

(Errors in n.m.

)

12 HKS 24 HRS 36 HRS 4 8 HRS 7 2 HRS

Official (NHC) 33 65 98 122 154

(No. of Cases) ( 43 ) / /I 1 \

( 41

)

( 39 ) (37) ( 33 ) CLIPER 37 73 119 161 216

//ION

( 43 ) ( 41 ) ( 39 ) ( 37 ) ( 33 ) SANBAR 28 55 92 141 302

( 15

)

( 15

)

( 14 ) ( 13 ) ( 11

)

QLM 81 90 119 172 268

( 19

)

( 18) ( 17 ) ( 16 ) ( 14 ) NHC83 38 61 88 106 178

(42) (40) (38) (36) (32)

BAM 50 84 123 154 268

(17) (16) (15) (14) (13)

NHC83 Revised 50 61 91 119 149

(39) (37) (35) (33) (29)

Figure 1. "Best-fit" smoothed track of successive eye positions for Hurricane HUGO

at 12-hourly intervals, based on aircraft and satellite fixes (begins when HUGO

had already attained full hurricane status on 9/13/89)

.

12
Figure 2. NOAA visible GOES satellite

view of HUGO shortly after it became

full hurricane at 16 GMT, 9/13/89..

South America coast in lower left,

north is toward the top.

Figure 3. Same as Fig. 2 except HUGO had

intensified further, photo at 1630 GMT,

9/13/89.

Figure 4. Same as Fig. 3 except near time of NOAA aircraft's first

penetration into HUGO at time of maximum strength, 1600 GMT on 9/15/89.

13
Figure 5a. Composite radar reflectivity PPI analysis from NOAA P-3

aircraft flight into Hurricane HUGO centered on 1727 GMT, 9/15/89.

Highest reflectivities (proportional to rainfall rates) are in

southwestern eyewall and a few-major rainbands. Total spatial

domain = 240 x 240 km. Courtesy Frank Marks, HRD/AOML/NOAA.

14
rM 172816 Z HURR. HUGO 9E

Figure 5b. Radar-cross section of reflectivities through eye of Hurricane HUGO

(NW-SE section in Fig . 5a) at 1728 GMT, 9/15/89 derived from NOAA P-3 aircraft's

3 cm tail radar. Courtesy Frank iMarks, HRD/AOML/NOAA . Height scale in increments

of 2 km.

TIME (MIN)

Figure 6. Radial plots of one minute sustained winds at 1500 ft flight-level and

derived surface pressure (thin line) through Hurricane HUGO, from NOAA P-3 aircraft

flight on 9/15/89. Courtesy Frank Marks, HRD/AOML/NOAA.

15
Figure 7. Same as Fig. 4 except visible satellite photo at 1330 GMT, 9/16/89

Hurricane HUGO was approaching Leeward Islands.

Figure 8. Composite time-series plots of measured minimum sealevel pressures (bold-line) in Hurricane HUGO vs. eye radii (dotted

curve) determined from USAF and NOAA reconnaissance aircraft and

surface radar data, from 9/15-23/89. Courtesy Hugh Willoughby,

HRD/AOML/NOAA . Four major cyclical eye stages are identified, and

occasionally when eye curve appears to jump upward, there was a double-concentric eyewall structure for a short time.

16
67° +

65° + 19°

Atlantic Ocean

6e» +

SI. Thomas

Culebra St. John

Vieques 64° + ,e°

Caribbean Sea 0=^

St. Croix

+

PUERTO RICO a U.S. VIRGIN ISLANDS

+ 17°

0 50 100 km

Atlantic Ocean

67.5 67.0 66.5 66.0 Figure 9. Upper map shows Puerto Rico and adjacent-owned islands of Culebra and

Vieques, as well as U.S. Virgin Islands principally impacted by HUGO. Lower map

is enlarged for details of rivers and towns in Puerto Rico mentioned in text.

17
Figure 10. Same as Fig. 7, except

satellite view at 1930 GMT, 9/17/89. Figure 11. Mesoscale analysis of Hurricane

Note Tropical Storm "Iris" to SE of HUGO'S eye positions , derived from tracking

Hurricane HUGO. of San Juan NWS radar as storm approached

within range (250 km) during late afternoon

on 9/17/89.

Figures 12a, b. Sequence of photos at times indicated on 9/18/89

from 10 cm NWS radar at San Juan airport, showing Hurricane HUGO'S

rainbands and eye, as it approached along track shown in Fig. 11.

Range-circles are shown every 50 km and St. Croix is located at

122° azimuth/150 km range from San Juan.

18
Figures 12c, d. Sequence of photos at times indicated on 9/18/89

from 10 cm NWS radar at San Juan airport, showing Hurricane HUGO'S

rainbands and eye, as it approached along track shown in Fig. 11.

Range-circles are shown every 50 km and St. Croix is located at

122° azimuth/150 km range from San Juan.

_St. Croix

Christionsted 2355 UTC

Frederiks-ted

1—

r

'7^y /C^—flngu ilia

\^ Barthelerny

( min . ) ( max . )

(J

(-ZD(3T

(VJ (M

J L J L I Z1 Kffl

I

89091711 HUGO

17 16 N 63 80 N

225140 Z

Loner Fuselage

2-»0 X 2'»0 km Minds at 1-min intervals Hurricane ResQarch

Di u i s i on

1 lOHhi/flOI-lL

Figure 13a. Composite analysis of radar reflectivities flight track and one-minute

sustained winds at 10,000 ft flight altitude of NOAA P-3 aircraft, centered around

time 2251 GMT, 9/17/89 as HUGO was nearing St. Croix (upper left) . Wind symbol

convention is triangle=50kts: and^:ea . full barb=10kts. Courtesy Marks and Powell, HRD.

19
89031311

Rlcit= 19 63 H Rlonr 66 62 W

Loner Fi-ijelaqe 300 X 500 km petition'

19 7ir( 66 , 62 II

Figure 13b. Similar to 13a, except composite radar analysis ( without aircraft

winds) when HUGO had weakened somewhat north of Puerto Rico, centered on time

2050 GMT, 9/18/89.

Caribbean Sea

Hurricane HUGO Rainfall (Inches)

7 a.m. 9/17 to 7a.m. 9/19/89

Figure 14. Analysis of storm total rainfall over Puerto Rico from Hurricane HUGO,

isohyets in inches. .Compare with rivers shown in Fig. 9.

20
100

HURRICANE HUGO

WSFO San Juan, PR

September 18, 1989

2400 400 800 1200

GMT

1600 2000

Figure 15a. Time-plot of measured peak gust windspeeds during Hurricane HUGO'S

strongest period at NWS San Juan, P.R.' See text for details. Courtesy of R.D.

Marshall

.

120

HURRICANE HUGO

Roosevelt Roads, PR

September 18, 1989

0)

a> Q

DC

O

2400 400 800 1200

GMT

1600 2000

Figure 15b; Same as 15a, except for Roosevelt Roads NAS , P.R.

21
Watch and Warning Infornnation for Hugo

7 2

<50 —

^ 48 —

I

V

0

36 —

24

12 —

62 63

56

hJormol Watch

A 1 42

35

Normal

War niiig

Watch Warning

i St. Croix I —I Vieques Puerto Rico

Figure 16. Histogram of warning-lead times from official NWS advisories for

Hurricane HUGO in Puerto Rico and U.S. Virgin Islands. Courtesy D. Wernly,

NWS/OM.

Figure 17. Comparison of final, smoothed track for Hurricane HUGO (bold-line) with

official forecasts of future track issued at 12-hourly intervals by NHC, Miami.

Courtesy of M.B. Lawrence and R. Case, NHC/NWS

.

22
Figure 18a. Apparent microburst damage in 2 residential areas of

coast, Vieques.

23
Monitoring the Aerodynamic Performance of a

Classic Suspension Bridge

by

Harold R. Bosch

ABSTRACT

More than 50 years ago, a classic girderstiffened suspension bridge was constructed near

the small town of Sedgwick on the Atlantic coast

of Maine. Spanning Eggemoggin Reach near Penobscot Bay, the Deer Isle-Sedgwick Suspension

Bridge is similar in design and appearance to the ill-fated original Tacoma Narrows Bridge.

The inherent flexibility of the design and the

selection of a relatively poor aerodynamic cross section have resulted in a bridge which has

behaved quite dynamically throughout its many

years of service. Despite periodic structural

modifications aimed at increasing the

structure's stiffness and controlling the windinduced motions, the bridge continues to exhibit

some sensitivity to the forces of wind. In 198I,

the Federal Highway Administration and Maine

Department of Transportation jointly installed

an instrumentation system on the Deer Isle

Bridge to monitor and evaluate its aerodynamic

performance. This paper will provide details on the instrumentation, describe the process of

data reduction, present some preliminary

findings, and discuss future plans for this

ongoing research study.

KEYWORDS: Aerodynamics; bridge; dynamics;

long-span; measurements; suspension; wind.

1 . INTRODUCTION

Full scale measurements at bridge sites are an important part of the Federal Highway

Administration's (FHWA's) wind research program.

The wind data obtained provides detailed

information regarding prevailing conditions

(speeds, direction, angle-of-attack ) and

turbulence properties (intensity, spectrum,

scale) which is often not available from other

sources. The structural response data provides

information on the structure's dynamic

properties (frequencies, mode shapes, damping)

which can be compared with computed values. Wind

and structural measurements are often used in

the design of wind tunnel experiments and to evaluate the effectiveness of aerodynamic

retrofits. They can also provide a means for

determining the adequacy of experimental and

analytical techniques.

(Washington). New and improved instrumentation i systems are currently in place on the Luling ;| Cable-Stayed Bridge (Louisiana) and Deer :j Isle-Sedgwick Bridge (Maine). Through a

j cooperative arrangement with the Maine | Department of Transportation (MEDOT), FHWA is

| monitoring the aerodynamic performance of the ' Deer Isle Bridge in order to provide

recommendations regarding bridge rehabilitation.

This study is well underway and will be

described in more detail in what follows.

2. DEER ISLE-SEDGWICK BRIDGE

The Deer Isle-Sedgwick Bridge, illustrated in

Figure 1 , is a conventional suspension structure | located near Penobscot Bay on the coast of

Maine. Built in 1938, the bridge spans

Eggemoggen Reach and consists of a girder-stiffened deck similar in cross section

j to the original Tacoma Narrows Bridge. The main

span is 1080 ft (329m) in length with side spans

of i\8H ft (118m) and approach spans of 100 ft

(HOm) for a total bridge length of 2308 ft

ilO^m) . The structure is symmetrical with two towers and the roadway has a 6.5% grade to provide vertical navigational clearance of 85 ft

(26m) at midspan. The 2-lane deck consists of a

^4.5 in (11cm) concrete slab having a width of 20 J ft (6m). Stiffening girders are 6.5 ft (2m) deep , and spaced 23-5 ft (7m) apart.

3. BRIDGE SITE

The bridge site is situated in Hancock County

approximately 55 miles (88km) east of Augusta

and 35 miles (56kra) south of Bangor. The bridge

carries State Route 15 over Eggemoggen Reach and

connects Sedgwick on the mainland with Deer Isle

and Stonington on the Islands. The longitudinal

axis of the structure lies on a NE-SW ej orientation with the mainland being at the north '| end of the bridge and Little Deer Isle at the Sj south end. The countryside to the north consists -j of heavily forested, rolling hills varying in

elevation to a few hundred feet. To the south

and east are many forested, low-lying islands,

j Deer Isle being the largest, with the Atlantic

Ocean beyond. To the west there is Penobscot Bay

with a small mountain range beyond.

1: In 1970, the FHWA installed its first field wind

instrumentation system on the Newport, Rhode

Island Suspension Bridge in cooperation with the

Rhode Island Turnpike and Toll Bridge Authority.

Since that time, wind and response measurements

have been taken on the Sitka Harbor Cable-Stayed ^Research Structural Engineer

Bridge (Alaska), Perrine Memorial Bridge Federal Highway Administration

(Idaho), and Pasco-Kennewick Intercity Bridge McLean, Virginia 22101-2296 U.S.A.

24
li. INSTRUMENTATION

4 . 1 Sensors

A large array of instruments has been installed

on the Deer Isle Bridge to monitor the wind

environment and structural response to wind

loadings. The positions of sensors on the bridge

are indicated in Figure 2. Mean wind speed and

direction are measured by two skyvane sensors

mounted on the top and bottom of the north

bridge tower. The upper skyvane is 18 ft (5.5m)

above the tower top while the lower one is

located approximately 33 ft (10m) above the

water. Air temperature is measured by three

thermistor probes installed at the top and

bottom of the same tower and at the instrument

house north of the bridge. Six tri-axis

anemometers are used to monitor wind turbulence

quantities (speed, direction, intensity, scale)

at the elevation of the bridge span. These

anemometers are mounted on 12 ft (3m) outriggers

cantilevered from the east side of the

superstructure and spaced in a logarithmic

array. Each anemometer set is aligned with the

bridge so that the "U" component is normal to the longitudinal bridge axis (facing southeast),

"V" component is parallel to the bridge axis

(facing northeast), and "W" component points

upward. The tri-axis anemometers employed are

dynamic instruments which are capable of

tracking velocity fluctuations in the frequency

range of interest.

Bridge deck response is monitored by six pairs

of single-axis, servo accelerometers installed

along the north side span and main span.

Accelerometer stations on the side span are

located at midspan and the quarter point nearest

the north tower. For the main span, stations are

located at midspan, both quarter points, and the

eighth point nearest the north tower. At each

station, an accelerometer mounted inside a weathertight enclosure is clamped to the bottom

flange of each stiffening girder. The axis of

each accelerometer is oriented to point upward

to measure vertical motion. The pair of

accelerometers at each station define the

vertical displacement and rotation of the span

at that point. The whole array of stations are

used to determine mode shape and frequency for

the structure.

To define tower motion, three accelerometers are

used at the top cross girder of the north tower.

Two of these are mounted at the top of each

tower leg with their axes oriented to point

northeast. This defines bending and twisting of

the tower top. The third is pointed southeast to indicate tower sway.

^.2 Recorder

Data recording equipment as well as lab test

equipment, parts, and supplies are housed in a small instrument house located just off the

north end of the bridge. This building is

outfitted with heat, air conditioning.

telephone, security system, and motor-generator

(for conditioning incoming power). Analog

signals from all sensors located on the bridge

are routed to this instrument house via

multi-conductor, shielded cable installed on the

west side of the bridge. Approximately 20,000 ft

(6096m) of cable was required to instrument the

Deer Isle Bridge. To insulate the

instrumentation from lightning strikes,

protection devices were installed in each

accelerometer enclosure and at the input panel

in the instrunent house.

The analog signals from the bridge sensors are passed through signal conditioning and amplified

to provide maximum resolution in the recording

system. Anemometer gains were adjusted to accommodate wind velocities up to 100 mph

(I6lkm/h) and accelerometer gains adjusted to handle .25g. Amplified signals from the deck

anemometers and all accelerometers are low-pass

filtered at 10 Hz to avoid aliasing of the data.

All signals are then routed to the recorder.

The recorder is an automated data acquisition

system (DAS) equipped with three 11 -track

cartridge tape drives providing enough capacity

for 24 hours of recording. Figure 3 illustrates

the data path and control logic employed by the

data acquisition system. The DAS continuously

scans up to 64 channels of analog input at a rate of 20 Hz. Data recording does not begin

until either wind velocities or bridge

accelerations exceed preset levels for a selected period of time. Once initiated,

recording continues until signals drop below

these threshold values. This mode enables

unattended operation and optimizes use of

available data storage capacity. While

recording, the system continuously scans all

channels, converts the analog signals to digital

data, and stores this data on magnetic tape for

later processing. When one tape fills, the

recorder automatically switches to another drive

and continues without loss of data. Wind speed

and direction measured by the two tower

anemometers are also recorded on strip-chart

recorders. This recording is continuous and

documents the day-to-day wind conditions at the

site regardless of whether or not the DAS is

active

.

5. DATA REDUCTION

After wind and bridge response data has been

recorded on digital tape, the tapes are forwarded to Turner-Fairbank Highway Research

Center (TFHRC) for preprocessing and evaluation.

This data reduction is accomplished by FHWA

staff using various microcomputers. Data tapes

are first scanned for errors. Raw data is

converted to engineering units using sensor calibrations obtained in the field and channel

statistics such as minimum, maximum, mean, and

standard deviation are compiled for 10/20-minute

blocks of data. This information is used to evaluate system performance and to identify

events which warrant detailed study. Mean wind

25
speed and direction obtained from the skyvane

sensors on the tower are displayed in polar form

and evaluated to detect any trends in the

recorded data. Figure 4 is a polar plot of wind

activity at the top of the north bridge tower.

This graph is updated each time new information

is obtained. Hourly means are obtained from the

continuous chart recordings and plotted in

time-history fashion for each month. As demonstrated in Figure 5, this technique is a useful means of compressing large volumes of

data and enables identification of significant

storm events.

As events of interest are identified during the

preprocessing, copies of the data are made for more detailed analysis using mainframes. The

data processing sequence is illustrated in

Figure 6. This analysis is accomplished in three

stages. In the initial stage, raw data is

converted to engineering units and summarized

using program TAPSUM much the same as was done

during preprocessing. The intermediate stage

prepares anemometer data from selected events

for detailed analysis by applying response

corrections and rotating the data into the

desired coordinate system. This is accomplished

using program MASTER. The final stage of data

processing involves spectral analysis of the

time series data. In ANAL2, time series data is

first prepared for analysis by point averaging,

detrending, and tapering. Next, fast Fourier

transform techniques are employed to process

data from 1, 2, or 3 channels simultaneously.

This analysis provides statistics on all the

data as well as power spectral densities and

auto correlations for the individual series. In addition, co-spectral and quadrature spectral

densities, coherence, squared coherence, phase

difference, and lagged cross-correlations are available for each pair of series.

To determine bridge displacements, the

"conditioned" acceleration time histories are processed through program COMBIN. Here, the

signals from the accelerometer pair at each

bridge station are averaged to extract the

bending component of motion. Next, the signals

are differenced to extract the twist component.

The acceleration time histories for each sensor

as well as those for bending and twist at each

station are then processed through MDIASM, a sophisticated double integration routine, to obtain the associated displacements. Finally,

these results are input to BRPLOT, containing

various 2-D and 3-D plotting routines, for

evaluation of mode shapes and animation of

bridge motion.

6. RESULTS

Wind spectra have been obtained for selected

wind events and compared with three empirical

functions —one proposed by Panofsky and Dutton,

another by Simiu and Scanlan, and a third by

Ramsdell, Typical spectra are presented in

Figure 7 for the U-components of four deck level

anemometers. The wind was from the southeast

with a mean wind speed of mph (20.5m/s) and a turbulence intensity of about )0%. The measured

spectra compare quite well with the Ramsdell and

"P/D" models.

To study the influence of the structure's bluff

cross section on the turbulence characteristics

as the wind passes around the bridge, anemometer

3a was moved to the west side of the deck

opposite anemometer ^ and redesignated 3b.

Figure 8 illustrates typical results obtained

from undisturbed and disturbed air flows

measured by sensors ^ and 3b respectively. For

this event, wind was from the southeast with a mean speed of ^40 mph (19m/s). The spectra from

anemometer 3b, located in the wake of the

structure, demonstrate a marked increase in

energy at the higher frequencies.

The along-span coherency of the wind was computed using the logarithmic spacing of the 6 deck anemometers. These spacings range from 105 to 1050 ft (32 to 320m). Coherency results for a 13 mph (20m/s) quartering wind from the south

are presented in Figure 9. These plots exhibit

the characteristic decay with increasing

separation distance between sensors. Approximate

exponential functions are included for

comparison. Similar results are presented in

Figure 10 for a 11 mph (20.5m/s) perpendicular

wind from the southeast. The coherence appears

to decay more rapidly for this case.

A summary of wind measurements at deck level

based upon data from 7 tapes is presented in

Figure 11. The polar plot illustrates the

distribution of mean wind speeds and directions

relative to the bridge span. The maximum wind

speed included in this data is 10 mph (19m/s).

Above the wind distribution plot is a graphical

representation of the associated peak

accelerations measured at all deck stations.

Although not all possible directions and speeds

are represented, this figure clearly

demonstrates the pronounced bridge sensitivity

to perpendicular winds. The maximum acceleration

observed in this dataset was O.lg.

The affect of wind speed on bridge motion is

illustrated in Figure 12. These results are

based upon measurements at accelerometer 5 and

anemometer 5. On the first graph, peak

accelerations are plotted versus the U component

of wind. For the second and third, RMS

accelerations are plotted versus the W and V components, respectively. A strong relationship

between bridge accelerations and wind speed is

clearly evident. The significance of wind angle

on bridge response in presented in the last

graph. Here, it is apparent that the average

level of bridge motion increases as the wind

becomes more perpendicular to the structure.

Figure 13 is an example of a measured

acceleration record and the computed

displacement time history for accelerometer 9. The maximum displacement computed from the

dataset was about 1 in (10cm).

26
7. CONCLUSIONS

All data collected to date has been preprocessed

and summarized; however, only a small portion

has received detailed analysis and evaluation.

Therefore, the results presented above must be

considered preliminary. Wind and bridge response

measurements at the site are still underway and

"production" processing of the extensive

database has just begun. With this in mind, the

following observations may be made:

0 spectral functions computed from wind

data look reasonable,

0 computed spectral functions compare well

with the Ramsdell model,

o the structure's shape generates strong

"local" turbulence with increased energy

at higher frequencies,

0 bridge motion is more likely to occur when winds are perpendicular to the

span, and

0 bridge activity tends to increase

steadily as wind speed increases.

DEER ISLE - SEDGWICK BRIDGE

Sadgwiok

(EUvatlon)

130 ft 484 ft 1080 ft 484 ft

" (147.5m) (329. 2m) " (14/. bm)

-- - --

'

T n 1

'

(Plan) Figure 1. Bridge Plan And Elevation Views,

130 ft

24 ft (7.3m)
0

TOP OF NORTH TOWER Figure 3. Data Acquisition System Schematic.

Figure Polar Plot Of Recorded Wind Events.

Figure 6. Data Processing Sequence.

28
CO

U

w

a,

CO

<c

z.o

M

COu

2:

(-1 Q

O

LEGEND SPECTRA RAMSDELl P&D

"sis J

or; o

w

a,CO

Q

o

NONDIMENSIONAL FREQUENCY (N»Z/U)

ANEMOMETER 2

NONDIMENSIONAL FREQUENCY (N»Z/U)

ANEMOMETER 3

>*

E-" M

COz

u

a

J

•a; ceH

o

Cd

a.

COJ

o

n

COz

wn

Q

O

2

NONDIMENSIONAL FREQUENCY (N»Z/U)

ANEMOMETER 5

n

CO2

CjJ Q

<

DC

E-U

w

Ol, CO

cC2

o

o

2

o

'_EGEMD _SPECraA_ RAMSOai

'

P A; D si's

NONDIMENSIONAL FREQUENCY (N*Z/U)

ANEMOMETER 6

Mean Speed: 44 m/h (20.5 m/s); Direction 80'

Figure 7. Wind Spectral Density For Four Anemometers - U Components,

29
ca

4-> o

<DQ. CO

-10 -10 Frequency

Anemometer 3b U-Component

tocQ

CO

i4-> o

0)Q.

Frequency

Anemometer 4 U-Component

>.

4-> •H

toc

o

<uQ.

toc

<u o

CO

S-. o

O.

UODC

P » 0

Frequency

Anemometer 3b V-Component

Frequency

Anemometer 4 V-Component

>> H

toc

<D a

CO

(DQ, to

toc

<D O

CO

t-,

-4-> O

0)Q.

CO

Frequency

Anemometer 3b W-Component

Frequency

Anemometer 4 W-Component

Mean Wind Speed: 40 m/h (19 m/s); Direction 90'

Figure 8. Wind Spectral Density For Upwind And Downwind Anemometers.

30
U-Component

exp [-4(^)]

V-Component

exp [-3(J^)]

Non-Oimensional Frequency (^) Non-Dimensional Frequency ijp-)

Figure 9. Along-Span Coherency For A Ii3 mph (20m/s) Quartering Wind (133").

U-Component

exp .[-8(^)]

V-Corrponent

exp [-5(^1]

nD> Non-Dimensional Frequency {jj-} Non-Dimensional Frequency (^)

Figure 10. Along-Span Coherency For A iJiJ mph (20ni/s) Perpendicular Wind (80")

Figure 11. Summary Plot Of Wind Speeds And Maximum Bridge Accelerations.

31
HPf 07

I«Pf 15 t«Pf !1 TAPE !7 ru^ Si

TAPE «

WIND SPEED (MPH) WIND SPEED (MPH)

a. MAXIMUM AND MINIMUM ACCELERATIONS DUE TO U-COMPONENTS c. STANDARD DEVIATION OF ACCELERATION DUE TO V-COMPONENTS

O lAPt 07

. lAPt 15 A I*PE 21 + TAPE ?7 0 lAPE J5 1 lAPE 37 0 TAPE 40

1-^

O TAPE 07

f TAPE 15 A TAPE 21 + TAPE 27 Q TAPE 35 a TAPE 37 0 TAPE 40

WIND SPEED (MPH)

b. STANDARD DEVIATION OF ACCELERATION DUE TO W-COMPONENTS d

Figure 12. Bridge Acceleration Versus Wind Speed And Angle - Accelerometer 5

ANGLE (degrees) STANDARD DEVIATION OF ACCELERATION DUE TO WIND ANGLES

Figure 13. Bridge Acceleration, Velocity, Displacement - Accelerometer 9-

32
Cooperative Research Program on Wind Tunnel

Testing in Turbulent Flow

BY

Koichi Yokoyama*', Robert L, Wardlaw*^, Hiroshi Sato*^/

Takaaki Kusakabe*'*/ and Michael G. Savage***

ABSTRACT

National Research Council Canada

and Public Works Research Institute/ Japan

are carrying out a cooperative research on

the wind tunnel testing method In turbulent

flow. This project includes spring mounted

rigid model test/ taut strip model test, and 2 phases full model wind tunnel tests using

heavy and light models.

In this report/ the outline of this research is explained at first. And summary of wind tunnel tests are introduced.

KEYWORDS

Wind tunnel/ Turbulent Flow/ Cable stayed

bridge/ full model test/ Cooperative research

1. INTRODUCTION

The design of long span bridges

requires the consideration of the effects of strong wind. Therefore/ wind tunnel test is usually performed to determine the

wind-induced response of the bridge at design

stage.

Many long span bridges are planned and constructed and the wind effects

on structures are eagerly studied in North

America/ Europe/ Japan and other countries.

However/ the wind tunnel testing method has

been developed in each country and the

standard testing method and assessment of the

results have not been established.

The Governments of Canada and

Japan have the technical cooperation program

on wind and seismic effects on structures.

National Research Council Canada (^JRCC) and

Public Works Research Institute (PWRI) are the counterpart organizations.

Recently/ both laboratories have

started a cooperative research project and

have jointly undertaken a series of winds

tunnel tests on wind-induced oscillations of cable-stayed bridge in smooth flow and in turbulent flow. Included in the project are a study on comparison of results of wind tunnel

tests performed in different wind tunnel

facilities and proposal on the standard wind

tunnel testing method of cable-stayed

br i dges.

In this report/ described are the outlines of the cooperative research work

between Canada and Japan and results of wind

tunnel tests consisting of spring mounted

rigid model test/ taut strip model test/ and

full model test.

2. OUTLINE OF THE PROJECT

2. I Purpose of the project

The purpose of this project is to establish procedures and guidelines for wind

tunnel tests of full bridge models in smooth

flow and boundary layer turbulent flow. This

will be performed by making the influence

derived from difference of method clear/

through comparing wind tunnel tests between

Canada and Japan.

The methods of wind tunnel tests have been developed in each country. There

have been exchanges of information on techniques of wind tunnel testing/ however

there are many differences among countries. So it seems significant to make the influence

derived from differences of method clear / through comparing wind tunnel tests between

Canada and Japan.

2.2 Program of the project

This project began in 1987 and

will be over in 1990. This research consists of spring mounted rigid model test/ taut strip

model test and 2 phases full model tests as shown in Figure 1. (1) Spring mounted rigid model test The spring mounted rigid model

test in smooth flow is carried out to study

fundamental characteristics of the bridge

deck. Results of this test are compared with

results of full model test. This test is performed only in PWRI.

(2) Taut strip model test The taut strip model test in smooth flow and in turbulent flow is carried

out to study fundamental 3-dimensi ona I effects and effects of turbulence on bridge

*l Head Structure Div. Public Works Research

Institute/ Ministry of Construction/ 1/ Asahi/

Tukuba-ShI/ Ibaraki-Ken/ 305/ Japan

*2 Head/ Low Speed Aerodynamics Laboratory/

National Research Council Canada/ Montreal

Road/ Ottawa/ Ontario, Canada, KIA/ 0R6

*3 Dr. Eng./ Head, planning Div. /PWRI *4 Research Engineer/ Structure Div./ PWRI

*5 Research Officer, Low Speed Aerodynamics

Lab./ NRCC

33
deck. Results of this test are also compared

with results of full model test. This test is performed only in PWRI.

(3)Ful I model tests. These tests are the core of this cooperative research. Similar full model tests were performed in both NRCC and PWRL and

they are compared. The final result of this cooperative research is drawn out from these

test. There are two phases of full model

tests. Namely/ heavy models are used in the

1st phase and light models are used in the 2nd

phase. Japanese research engineers joined

both phases of full model tests carried out in Canada and Canadian research officer joined on the 1st phase test in Japan.

At present/ all wind tunnel tests mentioned above have been completed.

Through discussing on these results/ it will become possible to establish manual or guideline on wind tunnel testing method.

3. DESIGN OF EXPERIMENT

3. I Bridge mode l for wind tunnel test

3.1.1 Deck Section

The shape was not intended to resemble any particular bridge but rather it was considered representative of modern plate

and girder construction. Figure 2 shows the

shape used in the full model tests. In NRCC/

there is full model of the Quincy Bridge. The

Ouincy Bridge model was built with a length

scale of 1:75 so that the overall length of 7220 mm would fit between the concrete fillets

of 9 m wide NRCC low speed wind tunnel. It seems that using this model is efficient and

economical. So in the 1st phase of full model

test/ the Quincy Bridge is referenced to decide characteristics of deck. Table 1 summarizes the deck sectional

characteristics.

The full model of PWRI has half geometrical scale. This scale was decided In consideration of the boundary layer wind

tunnel in PWRI which is 6 m wide and 3 m high.

All characteristics of the full model In PWRI are derived from the full model of NRCC

3.1.2 Cable and Tower Assembly

The deck was divided into three

spans/ in NRCC/ a center span of 3558 mm and

two side spans of 1781 mm- and was supported

by 56 cables attached to two towers. Figure 3

is elevation views of the installations and

shows the structural supports for the models.

In NRCC the towers and end points of the

bridge were supported on an isolated corner foundation to minimize the interference

created by the vibrations of the wind tunnel

shell/ but in PWRI the model was only set on

the floor of the wind tunnel.

The cables for the model were made from small diameter stainless steel piano

wire except the light model at PWRI in the 2nd

phase. In PWRL the geometrical scale is half

of NRCC's model/ and diameter of cables were

reduced in the 2nd phase of full model test. So it becomes impossible to get a piano wire

that is small enough. Coil springs were used

to adjust the elongation stiffness of cables

in the 2nd phase at PWRI. The stiffness of the

different cables were not constant and were

based on the requirements of the Quincy

Bridge model. The properties of the cables

are detailed in Figure 4. The cables were

anchored directly to the stiffening spines of the deck to minimize the mechanical damping of the bridge model.

3.1.3 Models for Spring mounted rigid model

test and taut strip model test.

Spring mounted rigid model test and taut strip model test are performed

before full model tests to predict the

fundamental characteristics of response*

which are determined by shape of cross

section/ in wind. Table-2 shows the condition

oft est s. The shape used In these tests did not have the stiffening spines required for the full model tests. Since the spineswere set

in the corners/ it was expected that their

presence would not significantly alter the

flow field around the deck. The major

difference between the deck sections of these

tests and full bridge model tests was the

higher reduced mass and inertia of the full bridge.

3.2 Flow properties

In regards to the full model test/

the bridge was tested in three different flow

conditions/ smooth flow/ and with moderate

and heavy turbulence. The summery of the flow

properties is given in Table 3. The moderate

turbulence was created with triangular spires

equally spaced across the inlet to the test section. The heavy turbulence was created by

mounting a flow trip across the test section

floor downstream of the spires. The smooth

flow condition has residual turbulence level. Figure 5 shows the arrangement of these

turbulent generators at PWRI for example.

The length scales for the

vertical and streamwise turbulence were based

on visual best fits with the Von Karman

spectr a. The properties for full model

test were not decided when the taut strip

model test was performed/ so that the flow

properties on taut strip model test were

different from full model test. Three kinds of turbulent flow were used for the taut strip

model t est

.

34
3. 3 Instrumentation of wind response

The motion of the bridge was monitored by the displacement transducers.

Displacement transducers are different

between NRCC and PWRI. In NRCC- transducers

are such types that are applied to an eddy

current. In PWRI they are such types that are applied to the photoelectric effect. So in NRCC/ targets are aluminum plates attached to the bottom of girder of model and pickups of transducers are put close to the targets. On

the other hand in PWRI/ targets are small

light bulbs and pickups of transducers are

put out of wind tunnel. Vertical and torsional

displacements are measured at center and

quarter of main span/ and center of side span. The wind velocity and turbulence were monitored by two ways. The turbulence of the wind simulation was monitored with a two channel hot film or wire probe. The mean

velocity was measured with a pitot tube.

torsional vibration. On vertical vibration/

the 1st peak (observed at the low wind

velocity) was the 1st bending mode/ and the

2nd peak (observed at the higher wind

velocity) was a high frequency bending mode.

The reduced velocity was about 1.0 on the 1st peak. On torsional vibration/ that was 1.2. There is good agreement between spring

mounted rigid model test and taut strip model

test except on the 2nd peak of bending

vibration.

In turbulent f I OW/ each of vortex-induced vibrations vanished or was reduced without changing of wind velocity.

(3)Flutter

The maximum torsional amplitude

becomes one degree at reduced velocity 2.5. This was not changed in turbulent flow so much as shown In Figure 8.

4.3 Full model tests In the 1st phase

4.3.1 The full model test in NRCC (1988)

4. RESULTS OF TEST

Analysis of results of wind tunnel

test is almost completed on spring mounted

test/ taut strip model test/ and the 1st phase

full model tests using heavy model. These

results are Introduced here.

4. 1 Spring mounted rigi d mod el test.

The results of the test can be expressed clearly in Figure 6 that shows

relation between wind velocity and amplitude

of vibration of a model. There are two wind

velocities which the amplitude of vertical

vortex-induced vibration become the peak

value. The lower reduced wind velocity Is about I and the higher Is about 2. Here/ wind

velocity is reduced or non-dimensional wind

velocity that Is amount of wind velocity

divided by the frequency of model and width

of deck. In regards to torsional vibration/

vortex-Induced vibration Is caused at about

1.2 and flutter Is caused at above 2.5 of reduced wind velocity.

4.2 Taut strip mode l t est

Relations between wind velocity

and response of model in smooth flow and in turbulent flow are shown in Figure 7 and

Figure 8 respectively. Where the intensity of streamwise turbulence lu is 5.3%. Results are summarized as written below

( 1) Buffeting

As comparing Figure 8 v/ith Figure

7/ buffeting Is caused by turbulence.

(2) Vortex-induced vibration

In smooth flow/ there are two peaks of wind velocity on vertical bending

vibration and there Is one peak on the

(1) Buffeting Response

The displacement recorded by the

three transducers at each station were

resolved Into vertical/ lateral/ and torsional

motion. The mean/ peak/ and standard

deviation of the responses for each of these

degrees of freedom were calculated for the

three valid stations for deck sections for all three flow conditions. The peak values are

the magnitude of the maximum deviation In either positive or negative direction from the mean position of the bridge. Figure 9 shows

the standard deviation and peak values as a function of the reduced wind velocity for the

torsional and vertical motions. For these

figures the velocity was reduced by the

appropriate deck section width and the

corresponding natural frequency In the first torsional or vertical bending mode.

A frequency analysis of the center span motion for the vertical and

torsional motion was carried out In both the

moderate and highly turbulent flow. The

dominant modes were the lowest modes on each

vertical and torsional vibration.

(2) Vortex-induced vibration

Vortex-induced vibration was not obser ved.

(3) Flutter

The test in smooth flow shows

torsional instability at reduced wind velocity

of 3.2. The RMS amplitude becomes 0.5 degree

at reduced velocity of 3.6.

4.3.2 The full model test In PWRI (1988)

(1) Buffeting Response

There Is no important difference

on Instrumentation and analysis at PWRI from

that of NRCC. Figure 10 shows the standard

deviation as a function of the reduced wind

35
velocity for the torsional and vertical mot i ons. According to frequency analysis/

the dominant modes were the lowest modes on each vertical and torsional vibration,

(2) Vortex-Induced vibration

Vortex-induced vibration was not obser ved.

(3) Flutter

The RMS amplitude becomes 0.5 degree at reduced velocity of 5.7.

5. DISCUSSION

Flutter and buffeting were observed in every type of test. In regards to vortex-induced vibration, it was observed in the spring mounted rigid model test and the

taut strip model test/ however it was not observed in the 1st phase full model test. The reason is a heavy mass of the full model. The

reduced mass of the full model is 3.6 and 4.4 times heavier than that of the spring mounted

rigid model and the taut strip model

respectively. Taking account of these results, mass and mass moment of inertia were reduced

to compare vortex-induced vibration in the

2n'd phase full model test. Also full models of the 1st phase were too heavy to compare

buffeting and flutter between full model test and spring mounted rigid model test or taut strip model test. Due to such reasons, discussion is limited to comparison of the 1st phase full model tests performed in NRCC and

PWRI. Further discussion will be performed

after completion of the analysis on the 2nd

phase of the full model tests.

5. I Similarity of test condition

It is very difficult to make

exactly similar flows and models. As a matter

of fact/ any wind tunnel tests have error and

restriction of facilities/ and the influence

they cause must be estimated. Due to this reason/ similarity of the test is compared

be I ow. Flow properties are summarized in Table 3. On high turbulent flow/ intensity

measured in PWRI was a little smaller than in NRCC. On length scale, there are considerable

and unexpected differences between NRCC and

PWRI. As intensity of turbulence affects the

amplitude of buffeting, making a correction

will be needed. On length scale, it is not clear how it influences. But the unexpected

difference itself must be further discussed.

In regards to characteristics of models. Table 4 summarizes differences. Only

structural damping becomes problem and

others were well adjusted. The difference of damping affects behavior of flutter

developing and onset wind velocity of flutter.

5.2 Flutter

Torsional responses in smooth

flow are compared in Figure 11. In NRCC,

flutter was caused at lower wind velocity than

in PWRI. When amplitude is small, one reason is supposed difference of damping. Figure 12 shows the influence of structural damping and

inertia obtained from spring mounted rigid

model tests in PWRI. As shown in Figure 12/ onset velocity of flutter increases with

damping and inertia increasing. The products

of structural damping and dimensi onl ess inertia become 0.10 and 0.64 in NRCC and PWRI respectively. These data suggest that it is not a strange difference on small amplitude

vibration. On the other hand/ there is a small difference of structural damping where

amplitude is large. But there is also not a small difference on large amplitude of flutter. There is no idea of the reason that explains this difference yet.

5.3 Buffeting

Amplitude of buffeting is compared in Figure 13. These data are measured at a station of center of the mair

span in the flow with high turbulence. It is clear that amplitude observed in NRCC is almost 2 times larger than in PWRI at every

wind velocity. The amplitude of buffeting is changed by intensity of turbulence. However

it seems there is not enough difference of intensity that causes such a difference of amplitude between by NRCC and by PWRI.

6. Concluding Remarks

In this report/ outline of the

project is explained and results of tests performed in the first 2 years are introduced.

Further discussion is needed to get final results. There is no comparison on vortex-induced vibration. On this problem,

the 2nd phase full model tests using lighter

models have been already completed in NRCC

and PWRI, and analysis is being performed.

There are no small differences

between the results of NRCC and PWRI. Of course, some differences should be inevitable.

But by analysing components and the reason of differences, many problems will become

clearer on the precision of the wind tunnel

test, methods of results correction, and

reasonable procedure for wind tunnel test. To

get these fruits/ not only advanced discussion

but also additional test will be needed.

ACKNOWLEDGEMENT

Authors wish to express

appreciation to the Japanese Science And

Technology Agency for financial aid to this cooperative research.

36
Fundanental Tests (1987)

Spring Mounted Rigid

Model Test

Taut Strip Model Test

Full Model Tests

The 1st Phase

Heavy Models

NRCC NRCC

(1988)

(1988)

The 2nd Ph: Liight Models

(1990)

PWRI PWRI

(1989)

Figure-1 PROGRAM OF COOPERATIVE WIND TUNNEL TEST

(l)NRCC

r 1

1

i 8 8.8

(2)PWR1

1 O)

:ST1FFENING SPINES

Figure-2 CROSS-SECTION OF BRIDGE DECK

Table-1 Deck Sectional Properties

Width (m) 0.177

Depth (in) 0.0177

Mass/Unit Length(Kg/ni) 3.31

Reduced Mass(in/pB^) 87

Mass Moment of Inert ia(Kg • m) 0.0174

Reduced Inertiad/^oB") 14.5

37
TRANSDUCER STATION i

(l)NRCC

(UN it: mm)

890.5

1 * 50

1829 890.5

(2)PWRI

CABLE NO ®

Figure-3 ELEVATION VIEW OF MODEL

EA(IO'N) CABLE lENSION(N) CABLE NO. EA(IO'N) CABLE TENSION(N) ® 6.2 3.81 ® 5.2 4.63 ®

®

®

5.2

8.6

8.8

5.52

6.78

8.22 ® 8.8 8.74 ® 10.8 9. 12

(UNlJi mm)

Figure-4 CABLE ARRANGEMENT (NRCC)

38
Table-2 CONDITION OF TEST

(SPRING MOUNTED TEST AND TAUT STRIP MODEL TEST)

dond i t i on Sdp i ne Mniint pH T n t ?> t r i D M nrl p 1 R i e 1 d Mode 1 Tp<; t Tpqt

n 1 U III D \m/ n AO n 1 n

L/C(-' 111 U \m/ U . U 1 Lciigiii ui nuuc 1 \u\j n Q7 I 9n 1 . C<J piabo/uiiii Lciiei-ii nivixB/ni/ H . ( U

0U . 911 coo acUULcU inaoS \in/ y D J Ma^Q Mniipnt of Inprt in (Ve • ricioo I'lVlidIv \J I IIICI tlu vr\o Illy n 117 Reduced Inertia fm/z^R") 3.61 1 34

Natural Frequency (Hz)

Vertical Bending Mode 1.97 7.61

Torsional Mode 3.86 13.9

Stractural Damping

(Log Decrement)

Vertical Bending Mode 0.020 0.025

Torsional Mode 0.017 0.041

TabIe-3 Characteristics of Turbulent Flow

^""\Caracter ist ics

name of Flow

Intensity Scale

lu l:/b Lw/B

NRC(I988)

Flow A 0.099 0.067 1.9 1.2

Flow B 0.207 0. 148 3.0 0.9

PWR I (1988)

Flow I 0.092 0.052 6.0 0.7

Flow n 0.163 0. 104 4.0 1.1

27100

lest iiectlon 3100

Trip Spire .

r Pitot Tube

Position of Model Position of Model (Turbulent Floff) (Smooth Flow)

t t c

H 1 1

i \ H 1

1

3000

1 2000

1 7800

Figure-5 ARRANGEMENT OF TURBULENCE GENERATER (PffRl)

39
-o

O.Ud

-a

c

o

o.

O) 0.U2 OS

<3J O

O

X -o

in a=

<v "Oo3-~

0)

rac

o n

a.

V)<D OS

CSO

»-> t-

<D

5.U

CO

4.0

o

Q.

CO :t.u

oc;

C8= -2.0 o

COo

I.U

' ooooo ^/(f'B) U/(faB)

Figure-6 RESULTS OF SPRING MOUNTED TEST

O Haxitiut X RNS O HiniuuB

Q.

to

CO

CP

O HaxluuB

O Minieuii

o

C X y 5 o 6 8 10

U / ( f ,B )

3 4 5 U / ( f , B ) Figure-7 RESULTS OF TAUT STRIP MODEL TEST (SMOOTH FLOW)

O MaxiauB X RHS O KiniaUEi

c

10

a.

COOJ OS

o

O MaxiBUB X RHS O KiniBUB

O X

o ^

o

U/(f,B) U/(feB)

Figure-8 RESULTS OF TAUT STRIP MODEL TEST (LOW TURBULENCE)

40
Z 0.07|

I 0.06

I

0.05

i 0.04

- 0.03+

I

0.02

t 0.01- " 0.00

-0.01

0

^

0.07-

I

0.06-

I

0.05

i 0.04-

E 0.03

OS- 0.02- 03 ~ 0.01

i 0.00

-0.01

0

MAXIMUM

RMS

MAXIMUM

RMS

6 8 10

U / ( f , B )

12

o 1.0

5 0.5-

0.0

(1) In Smooth Fl ow

MAXIMUM

RMS

^ 3.0t

-o^ o t;

CO C»\J

4- -4 6 8 10 12

U / ( f , B )

I 2.0

OS 1.5-

co

« 1.0

o" 0.5 i

0.0

MAXIMUM

RMS

0

(2)ln Heavy Turbulent Flow

Figure-9 RESULTS OF FULL MODEL TEST (NRCC)

41
^ ,05

^ ^ ( f eB) U/(f«B)

(1) In Smooth Flow (2)|n Heavy Turbulent Flow

Flgure-lO RESULTS OF FULL MODEL TEST (PWRI)

I

42
Table-4 COMPARISON OF MODELS BETWEEN NRCC AND PWR[

bona 1 1 1 on PWR 1 r n i\ I

niain d \mj U . I / / 0 u. uooo uepm u \n\j 1 n n 1 7R u . U i / 0 U . UUOC)

LcllgLll Ul lilUUcl L VlTl/ Rl ncUUucU ludao \m/ y D ) OD . o

RoHiir>orl Inortia (ml n^^\ 1

1

11 . 71 liaiuiai ricmjciiuy \\\L

)

icl L 1 La 1 DullUlilg I'lUUc Torsional Mode 8.59 14.4

Stractural Damping

(Log Decrement)

Vertical Bending Mode

Measured at amplitude of B/200 0.018 0.024

Torsional Mode

Measured at amplitude of 0.5° 0.055 0.041

Measured at amplitude of O.T 0.007 0.055

. 0:PtfRI

A :NRCC

0

A.

o

o

Figure-11 COMPARISON OF FLUTTER

43
0.02

O : PWRI

A :NRCC

0.01

A

O

o

o

o

o

o

a\(^0 9

o o

-I L.

1 0

U / ( f ,B )

Figure-13 COMPARISON OF BUFFETING

44
Wind on Offshore Platforms

by

Charles E. Smith*

ABSTRACT

Increasing political and economic demands for more

energy have focused a need to develop technologies

to explore and produce oil and gas resources

offshore. These technologies include methods to analyze and describe data on natural events like

wind, waves, currents, ice, and earthquakes, and

to compute the loads such events produce on offshore facilities. This paper reviews the

currently accepted procedure for determining wind

forces on offshore platforms on the U.S. Outer

Continental Shelf (OCS). It also includes a brief

overview of the relationship between wind and other

environmental events that exert loads on offshore

structures.

KEYWORDS: Offshore platforms; oil and gas

operations; wind criteria; wind loads.

.1. INTRODUCTION

The fundamental and practical aspects of modern

wind engineering practice as it relates to design

of offshore facilities have evolved from methods

developed for land-based structures. This

evolution is the result of a more directed

awareness of wind-related problems, the trend for

higher and more flexible buildings, and the

increasing use of cladding systems for which wind

loading is the primary design consideration.'

Wind loads are not unique for offshore structures,

however, over water-winds tend to be more severe

because of the lack of sheltering effects of

terrain, and the fact that storm systems gain more

energy over water. What is unique for offshore

structures is that maximum or near-maximum wind

load may occur simultaneously with maximum wave and

current loads and higher than normal water levels.

The major mass and resulting wind force area of an offshore facility is located at the uppermost

elevation of the structure so that the wind loads

are applied at levels that produce maximum overturning moments.

2. EFFECTS OF WIND

What is wind? This may be a trivial question to some, but a brief description may provide a better

perspective of its relation to offshore structures.

In essence, wind is the horizontal movement of air

in response to differences in atmospheric

pressures. The earth is traversed by major surface

wind systems known as trade winds which interact

with the water of the oceans to produce major

current systems. Within the major wind systems,

short-term disturbances occur, and are responsible

for the extreme winds of interest to the structural

engineer when he designs offshore facilities.

The disturbances that are of primary importance to the design of offshore structures in the United

States are those associated with tropical

hurricanes. Understanding the basic mechanics of

these storms, and the way they relate to the

development of other loads on the structure is very

important. The effect of wind may manifest itself

in several ways; wind loads that act directly on structures, wind generator waves, wind generator

currents, and wind generator surge conditions

(fig. 1).

For offshore facilities, wind forces act upon the

portion of structures exposed above the water

level. This includes any equipment, deck houses,

and derricks which are located on the platform as shown in figure 2. The major parameters used for

design purposes are as follows:"

1. the frequency of occurrence of sustained

wind speed from various directions,

2. the spatial variations of the basic wind

speed, in particular, the variation with

altitude, and

3. the temporal variations in the basic wind

velocity manifested as gusts.

For design, the wind speed may be classified as gusts, which are wind velocities that are averaged

over less than 1 minute, and sustained winds which

Research Program Manager

Technology Assessment and Research Branch

Minerals Management Service

Herndon, Virginia 22070-4817

45
are velocities that are averaged over 1 minute or

longer. Most engineers recommend that wind data

be adjusted to a standard elevation, such as 33

feet (10 meters) above mean water level, using a specified averaging time such as 1 hour. Wind data

can be adjusted to any specified averaging time or elevation using standard profiles and gust factors

as is shown in a later section of this paper.

Building codes have traditionally specified the

wind pressures to be used in the design of

buildings and other land-based structures.

However, since offshore structures are more susceptible to extreme winds than land-based

structures, it is much more important that design

criteria accurately assess maximum loading events

than those specified by building codes. Reference

2 provides guidelines for extreme wind speeds for

several areas in U.S. waters. To supplement this

information, hindcasting is considered the most

appropriate technique for developing the necessary

environmental exposures (fig. 3).'*

The Minerals Management Service (MMS) requires that

offshore structures be designed to withstand

loading events, both wind and wave, with return

periods of 100 years unless the owners can demonstrate, through appropriate analyses, that

less severe design events are appropriate.' For

operations elsewhere, particularly in more harsh

environments where new platform concepts are proposed, different design criteria may be

required. For example, in the deep-water regions

of the Gulf of Mexico, some companies use a 200-

year return period.

3. OTHER WIND EFFECTS

As opposed to land-based structures, winds offshore

are primary contributors to other environmental

events which exert loads on offshore structures.

Wind-driven waves, produced by the friction that

exists between the wind and the ocean surface have

the single greatest effect on offshore operations.

Wind-driven waves develop and grow when a certain

wind prevails from the same direction for a period

of time. The length of the unrestricted path of wave development is called the fetch as shown in

figure 4. The fetch can also be defined as the

distance over which the wind blows to generate

observed waves at a given position (fig. 5). Four

factors, wind speed, duration, fetch, and water

depth determine wave height (fig. 6). For example,

if a 30-knot wind blew over an average of water of

unlimited fetch for 5 hours, 7 1/2-foot waves would

develop. If the same 30-knot wind blew over the same area of water for 30 hours, the waves would

reach 16 feet.*

In addition to wind-driven waves, tides, and winddriven currents are also important considerations

in platform design. In determining the maximum

water level for platform design, the storm-tide

elevation is the datum upon which storm waves are superimposed. Wind-generated currents are combined

in a vector sum with tidal currents (associated

with astronomical tides) and circulational currents

(associated with oceanic-scale circulation

patterns). Wind-driven currents can add

substantially to the loads on offshore structures

because the loading function (Morrison equation),

is based on the square of the water particle^

velocity. «

4. General Design Criteria i

General requirements for determining wind loads on offshore oil and gas facilities on the U.S. OCS are contained in the American Petroleum

Institute Recommended Practice 2A (API-RP2A). This,

document recommends that sustained wind velocities

be used in the computation of overall platform wind'

loads and that gust velocities be used for the!

design of individual structural elements (fig. 7). |

The variation of wind velocity with respect to timel

is shown in figure 8 and the variation with heighti

is given by the relationship as shown in figure 9.:

The referenced wind velocity height is taken at 33:

feet (10 meters) above the referenced water depth.

The exponent used in the velocity relationship is

usually assumed to be between 1/7 and 1/13'

depending upon the sea state, relative distance;

from land, and the duration of the design wind!

velocity. For most designs in the open ocean, the

exponent is approximately equal to 1/13 for gusts j and 1/8 for sustained winds. Typical velocity'

coefficients for various wind durations and heights i are given in figure 10. The API-RP2A document

states that the wind force on a structural member

should be calculated by using the following wind

velocity/force relationship: -

i: F = 0.00256 (V)' CA

Where: i

F = Wind force (lbs)

V = Wind velocity (mph)

C = Shape coefficient

A = Projected area (ft^)

As in most onshore designs, local wind effects such

as pressure concentrations and internal pressures^

must be considered. For these situations, it isyj

recommended that the same analytical guidelines be i used as set forth in Section 6 of the ANSI A58.1-

82 Standard, "Building Code Requirements fori

Minimum Design Loads in Buildings and Other

Structures." In general, for all wind approach

angles to the structure, forces on vertical

surfaces are assumed to act normal to the surface

and in the direction of the wind. For surfaces not

in a vertical attitude with respect to wind

direction, appropriate relationships should be used

to take into account the direction of the wind in

relation to the attitude of individual members

(fig. 11).

In the absence of more refined data, API-RP2A

recommends the following shape coefficients (C) for

objects perpendicular to wind flow: 1

I Beams 1.5 I. Sides of Buildings 1.5 ; Cylindrical Sections 0.5 i

46
Overall Projected Area of

the Structure 1.0

The question of shielding is left to the judgment

of the designer. No specific guidelines are set forth; however, if the second object is considered

to be located close enough behind the first to warrant the use of a shielding coefficient to limit

the load on that object, then it can be used.

Typical shielding values for structural members are given in reference 7.

Certain requirements are stated for so-called windsensitive structural components. These components

are defined as equipment or objects located on the

platform that are exposed to wind and whose height

exceeds five times their horizontal dimension, or whose dynamic properties make them very responsive

to the action of the wind. For these objects,

detailed analyses are required where the timevarying aspects of the wind loads acting on the

object must be considered. It is recommended that

the analytical procedure, as set forth in Appendix

A6 of the ANSI A58.1-82 Standard, be used as a guide in determining gust loadings on windsensitive appendages on a platform.

5, Other Wind/Force Criteria

In addition to the API recommended method, the

American Bureau of Shipping (ABS) and Det norske

Veritas (DnV) have also established formulas to

calculate wind forces on offshore structures. Of

the three, the DnV formula provides the most

conservative values, whereas the ABS formula yields

values slightly less than those obtained using the

API method. Reference 8 contains a useful

comparison between the methods with a detailed

explanation of the coefficients and their

variability with height above water level and wind

speed.

6. Compliant Structures

Compliant structures are more susceptible to the

dynamic effects of wind than are conventional fixed

leg platforms. The presence of waves, wind, and

currents causes these platforms. Because of their

compliance to have mean and fluctuating offsets in

the direction of the applied loads. Compliant

structures can have natural periods that are much

longer than those of conventional platforms.

It is well known that significant fluctuation of

wind velocity occurs at periods up to hundreds of

seconds. Figure 12 shows typical wind and wave

spectra with typical platform periods noted for

comparison purposes. From this figure, it is

easily seen that the wind can provide large energy

inputs for certain degrees-of- freedom of a tension

leg platform.

Many models of wind spectra have been proposed in

design codes and the scientific literature.

Unfortunately, they disagree in the low frequency

range which is of most interest for the design of

compliant structures. Much of the data upon which

these model spectra are based were obtained over

land, and measurements in strong winds,

particularly hurricanes, are scarce.'

Wind-tunnel tests have proved useful for estimating

the fluctuating wind loads acting on platforms.

The extent to which these fluctuating loads can cause resonant-type conditions in compliant

structures is very difficult to ascertain.

Reliable information on full-scale hydrodynamic

damping, which controls the magnitude of the

resonant amplifications, is not available.

Reference 10 presents various aspects of windtunnel testing of compliant structures. The

reference discusses various methodologies for

obtaining and using wind-tunnel data and techniques

for measuring wind and wave effects simultaneously.

7 . Concl usions

This paper was written to provide the reader with

an understanding of the key considerations in

determining the response and performance of

offshore facilities to wind-induced loads. As

compared to onshore structures, wind loads on offshore structures act in combination with other

loads produced by waves and currents. Wind loads

are not unique for offshore structures, but

because of their significant effects in combination

with the other environmental loads, they must be

carefully considered in the design process. This

is absolutely vital in the design of compliant

platforms. The references cited should be

consulted to acquire a more rigorous overview and

appreciation of the aspects of winds acting on offshore facilities.

8. References

1 . Wind Loading and Wind-Induced Structural

Response, A State-of-the Art Report . Prepared

by the Committee on Wind Effects, American

Society of Civil Engineers, New York, New

York, 1987.

2. Recommended Practice for Planning, Designing,

and Constructing Fixed Offshore Platforms , API Recommended Practice 2A {RP 2A), Eighteenth

Edition, American Petroleum Institute,

Washington, D.C., April 1, 1989.

3. Gaythwaite, J., The Marine Environment and

Structural Design , Van Nostrand Reinhold

Company, New York, New York, 1981.

4. Environmental Exposure and Design Criteria for

Offshore Oil and Gas Structures , Marine Board,

National Academy of Sciences, Washington,

D.C., May 1980.

5. U.S. Code of Federal Regulations , 30 CFR, Part

250--0il and Gas and Sulphur Operations in the

Outer Continental Shelf, Subpart I--Platforms

and Structures.

6. Wind, Waves, and Weather , Unit V, Lession 1,

Rotary Drilling, Petroleum Extension Service,

University of Texas, Austin, Texas 1984.

47
7. Wind Loading and Wind-Induced Structural

Response, A State-of-the-Art Report , Committee

on Dynamic Effects of the Structural Division,

American Society of Civil Engineers, New York,

New York, 1987.

8. Hsu, T. H., Applied Offshore Structural

Engineering . Gul f Publ ishing Company, Houston,

Texas, 1984.

9. Forristall, G. Z., Wind Spectra and Gust

Factors Over Water . OTC 5735, Offshore

Technology Conference, Houston, Texas 1988.

10. Wind Effects on Compliant Offshore Structures , Proceeding, Structural Congress '86, American

Society of Civil Engineers, New Orleans,

Louisiana, September 15-18, 1986.

(

/

/;

/

/

//

/ IMPORTANCE OF WINDS

Figure 2 Model offshore platform illustratinj

typical load sources. WIND LOADS

WAVE GENERATION

SURGE GENERATION

CURRENT GENERATION WIND DATA SOURCES

Figure 1 Examples of important wind contribution

to loads on offshore structures.

VISUAL OBSERVATIONS

COMPUTATIONS FROM PRESSURE

MEASUREMENTS

Figure 3 Typical means for determining win

speeds offshore.

48
OFF WEST COAST

Figure 4 Moving wind field illustrating wave

growth path towards a structure

at a fixed location.

Wave Growth and Propagation

Ripples Chop Sea or Swell Wind Waves

Fetch H

API - RP2A

1.3.2 Winds

2.3.1c Dynamic Wave Analysis

4. Winds

2.3.2 Wind

2.3.2a General

2.3.2b Wind Velocity Profiles and Gust Factors

2.3.2c Wind Velocity and Force Relationship

2. 3. 2d Local Wind Force Considerations

2.3.2e Shape Coefficients

2.3.2f Shielding Coefficients

2.3.2g Wind Sensitive Structures

2.3.2h Wind Tunnel Data

Table 2.3.4-2 Guideline Extreme Wind Speeds for Twenty Areas in the United States Waters

Figure 7 Section of API-RP2A which applies to wind aspects of platform design.

Figure 5 View of a wave generating area, showing

development of smaller, generally

steeper and short-crested waves in the

early stages.

WAVE GROWTH WITH TIME AND FETCH

140 KT WINDS)

UNLIMITED FETCH UNLIMITED TIME

WAVE

HEIGHT 20

(FT) 10

TIME VARIATION OF WIND SPEED

60

WIND SPEED

(MPH)

40

20

MAX GUST / MAX1-MINMEA H 10-MIN MEAN

1 1 1 1 1111 2 4 TIME (MINI

0 8 16 24 32 40 DURATION (HRS)

0 60 120 180 240 300 360 420 480 Figure 8 Variation of wind speed at a fixed

FETCHiN.Mi.) elevation showing typical averaging time

interval s.

Figure 5 Collation of wave height, wind duration,

and fetch.

49
WIND PROFILE

SPEED V

Figure 9 General wind profile recommended by API- RP2A in designing offshore platforms.

A,B

HEIGHT AVERAGING TIME INTERVAL ABOVE SWL 1 HR 10MIN 1 MIN 15 SEC 5 SEC 3 SEC

(M) (FT) A B C 0 E F

10 32.8 1.00 1.06 1.22 1.32 1.40 1.43 20 65.6 1.10 1.17 1.34 1.42 1.50 1.53 30 98.4 1.16 1.24 1.41 1.48 1.56 1.59 40 131.2 1.20 1.29 1.46 1.52 1.60 1.64 50 164.0 1.24 1.32 1.50 1.56 1.64 1.67 60 196.8 1.27 1.35 1.53 1.58 1.66 1.69 BO 262.4 1.31 1.40 1.57 1.63 1.71 1.74 100 328.0 1.34 1.43 1.62 1.66 1.74 1.77 120 393.6 1.37 1.46 1.65 1.69 1.77 1.80 140 459.2 1.39 1.48 1.67 1.71 1.79 1.82 150 492.0 1.40 1.50 1.69 1.72 1.80 1.83

TYPICAL AVERAGING TIME INTERVALS USED PGR ANALYZING AND REPORTING DATA

THE EXPOSED SUPERSTRUCTURE, REGARDLESS OF DIMENSIONS,

TO BE USED WHEN DETERMINING THE WIND FORCES ASSOCIATED

WITH A MAXIMUM WAVE OR CURRENT FORCE, WHICHEVER PREDOMINATES

PARTS OR THE WHOLE OF THE SUPERSTRUCTURE ABOVE THE LOWEST STILL WATER LEVEL WHOSE GREATEST DIMENSION, HORIZONTALLY OR VERTICALLY. EXCEEDS 50 METERS (164 FT), FOR WHICH A FIFTEEN SECOND MEAN SPEED APPLIES

Figure 11 Wind-exposed areas on a model

floating offshore drilling rig.

of 3

u

oa

w

0

O

o

Surge. Sway

& Yaw Periods Ot TLPs

Wind Spectrum

Jacket & Gravity Platforms

Heave, Pitcti & Roll Periods Of TLPs

Wave Spectrum

1000 100 10 -1

Period, sec.

E PARTS OR THE WHOLE OF THE SUPERSTRUCTURE ABOVE THE LOWEST STIl L WATER LEVEL WHOSE GREATEST DIMENSION, HORIZONTALLY OR VERTICALLY, DOES NOT EXCEED 50 METERS

(164 FT). FOR WHICH A FIVE SECOND MEAN SPEED APPLIES

F INDIVIDUAL MEMBERS, AND EQUIPMENT SECURED TO THEM ON OPEN DECKS FOR WHICH THE THREE SECOND MEAN GUST SPEED APPLIES

Figure 12 Excitation and frequency ranges of

offshore platforms.

Figure 10 Wind velocity factors for differing

average time intervals.

50
Theme II

Storm Sm:*ge and Tsmiamis
I

i:

;

B

a

t
Deep Ocean Recordings of Tsunamis

by

Frank I. Gonzalez^

ABSTRACT

Excellent deep ocean tsunami

measurements were recently acquired in

the Gulf of Alaska, using bottom

pressure recorders developed over the

last decade at NOAA's Pacific Marine

Environmental Laboratory. The

observations were made as part of the

Pacific Tsunami Observation Program,

which has maintained a deep ocean network since 1986. Successful

acquisition of these data demonstrates

the feasibility of long-term tsunami

monitoring in the deep sea with selfcontained instrumentation on the ocean

bottom.

KEYWORDS: Tsunami; pressure;

measurements; PacTOP

.

1 . INTRODUCTION

Along the U.S. and Hawaiian coast, six

major tsunamis have inflicted about 350

casualties and half a billion dollars

in property damage over the last 44

years (Bernard and Goulet, 1981) . Along the entire Pacific rim, an average of one destructive tsunami a year has occurred over the last 100

years, and more than five tsunamis are generated each year of sufficient

amplitude to be observed (Lockridge,

1987) . High quality measurements of even these smaller tsunamis would be

extremely valuable to tsunami research.

Indeed, such data are now essential to

further improvements in our understanding of tsunami generation and

propagation, since theoretical

developments and huge increases in

computational power over the last two

decades have far outstripped the

usefulness of our existing

observational data base.

This is because most tsunami

measurements in the data base have been

made by coastal tide gauges, which are

NOAA/Pacific Marine Environmental

Laboratory, 7600 Sand Point Way NE,

Seattle, WA 98115

inadequate for a number of reasons. Small tsunami amplitudes and periods

are typically an order of magnitude

smaller than the tidal signals for

which a tide gauge is designed; as a result, time and height scales of a tide gauge are generally inadequate.

The gauges are frequently characterized

by a nonlinear response and are invariably located in protected inlets

and harbors which may themselves have

complex responses to an incident

tsunami. Local atmospheric forcing can increase the background noise level in

the tsunami frequency band and severely

decrease the signal-to-noise ratio

(SNR) , so that tsunami arrival is

unambiguous only when it is

serendipitously preceded by a sea level

record which is quiet compared to the

incident tsunami energy. But the

arrival of the tsunami itself can then

excite local resonant modes of

oscillation

.

Any of these factors can seriously

distort the signal of interest and

render the records useless for research

into the fundamental physics of tsunami

evolution: generation, deep ocean

propagation, shoaling, and coastal

inundation. These investigations

require accurate tsunami measurements

in the deep ocean and at coastal sites

fully exposed to the open ocean.

2 . PacTOP

To acquire the necessary data, the

National Oceanic and Atmospheric

Administration (NOAA) and the U.S. Army

Corps of Engineers (USACE) have entered

into a collaborative effort known as the Pacific Tsunami Observational

Program (PacTOP) , which has developed

and maintains a long-term tsunami

monitoring network. NOAA is

responsible for the continued

occupation and maintenance of five deep

ocean stations, utilizing bottom

pressure recorders and moorings such as those shown in Figure 1; the USACE

provides coastal tsunami measurement

53
capabilities at six primary pressure

gauge sites in shallow water exposed to

the open ocean, and three secondary

sites situated in coastal inlets

(Gonzalez, et al., 1987).

The deep ocean BPR network is designed

to monitor critical portions of the

seismically active Aleutian-Alaskan

Seismic Zone. This region generated

three of the six most recent tsunamis

that attacked the U.S. and Hawaiian

coast; furthermore, these three

tsunamis accounted for more than 82% of

the casualties and 85% of the property

damage (Bernard and Goulet, 1981) . Future tsunami threats to Alaska, the

U.S. west coast and the Hawaiian

Islands are concentrated in sub-regions

of the Aleutian-Alaskan Seismic Zone

known as the Fox, Unimak and Shumagin

Seismic Gaps; these have been

identified with significant potential

for large tsunamigenic earthquakes

(Nishenko and Jacob, 1990)

.

One source station is located in the

Shumagin Gap itself, on the landward

slope of the Aleutian Trench, three

directional array stations form a triangle centered about 250 km

southeast of the source, and one farfield station is located 2200 km

distant from the source, near the

great-circle route from the Shumagin

Gap to Crescent City, California. Four

of these five station locations are shown in Figure 2; the fifth station is

located at the southern apex of the

equilateral triangle formed with

stations AK7 and AK8, and is not shown

because the BPR was not recovered at

this site during the 1987-88 field

season

.

3. INSTRUMENTATION

Two types of BPR units and their

associated moorings are presently in use (Figure 1) . Each BPR utilizes a Paroscient if ic Digiquartz Model 410K

(0-10,000 psia) transducer and records

a 56.25-second average pressure every

56.25 seconds (64 samples per hour),

with a resolution better than 1 mm

equivalent sea water head for signals

of period of a few minutes or longer.

(A new generation of BPRs currently

under development will be capable of

acquiring 15 s samples with the same resolution.) The units are entirely

self-contained and store data

internally, with a capacity

corresponding to about 14 months of

operation; this allows for early summer deployment and late summer recovery, if

necessary. Eble et al (1989) discuss

instrument characteristics and BPR data

processing procedures in some detail.

Schematics of the two mooring designs

are presented in Figure 1 . Important

ancillary components shown include an acoustic release that disengages an anchor, sufficient buoyancy in the form

of glass spheres or or syntactic foam

to raise the anchor-less unit to the

surface, and a marker buoy with a number of recovery aids (VHF

transmitter, strobe light, radar

reflector, marker flag)

.

Two or more oceanographic cruises are conducted each year to recover and re- deploy BPRs at each of the deep ocean PacTOP stations; as a practical matter.

Gulf of Alaska weather patterns

restrict these cruises to the summer months . Out of seventeen

deployment /recovery cycles to date, one instrument has been lost, another has

flooded, and some data loss has

occurred in the records obtained;

equipment and data recovery rates for

the project are currently 88% and 82%,

respectively

.

4. TSUNAMI RECORDINGS , -

'

Three large intraplate earthquakes

recently occurred in the Alaskan Bight,

with magnitudes 6.9 Mg on 17 November

1987, 7.6 Ms on 30 November 1987, and

7.6 Ms on 6 March 1988 (Lahr et al.,

1988) . The epicenters were somewhat

south of the subduction zone associated

with the Yakataga Seismic Gap, a region

similar to the Shumagin Seismic Gap in

that it is also associated with a high

probability of experiencing an eventual

great tsunamigenic earthquake (Savage

and Lisowski, 1986; Nishenko and Jacob,

1990) . , .

The transformation of seismic energy to

tsunami energy during all three

earthquakes was relatively inefficient,

since the apparent faulting mechanism

in all three cases was primarily

strike-slip, and most of the crustal

motion was therefore horizontal.

Nonetheless, each earthquake did

generate a small tsunami which was

54
I detected both on the coast and in the

deep ocean. Gonzalez et al. (1990a)

provide a discussion of all of the

PacTOP and coastal tide gauge

observations of these events, and

Gonzalez et al. (1990b) have compared

I the deep ocean BPR observations with

j numerical simulations using a nonlinear

shallow water tsunami model.

Figure 2 is a summary map for the

earthquake and tsunami of 6 March 1988.

Contours of a quadrupole feature are displayed southwest of Yakutat which

represent estimates computed using

Okada's (1985) expressions for the

ratio of vertical seafloor displacement

to total seismic slip; units are cm/m,

the contour interval is 1, outermost

contours have the values ±1, solid

lines indicate uplift, and dashed lines

indicate subsidence. Computed extrema

and rms values were ±12 cm/m and 3 .

8

f cm/m, respectively. Lisowski and

j Savage (1989) estimated total right-

,[ lateral slip due to both the 30

; November and 6 March earthquakes to be

2.9 ± 1.2 m using geodetic means, and

! 4.5 m using the seismic moment. The

average of these two estimates is 3.7

j m, and if half of this slip occurred on

6 March, then the model computations

suggest extrema and rms vertical

' displacements of ±22 cm and 7 cm,

I respectively.

j Tsunami records obtained at the PacTOP

j deep ocean stations and Alaskan coastal

'{ tide gauge stations are also displayed

; in Figure 2 . BPR data are 60-min highI pass filtered, and tide gauge data are

' 3-60 minute band-pass filtered,

j Tsunami estimated time of arrival

] (ETA), as computed by NOAA's Pacific

I Tsunami Warning Center, is indicated by

;i a vertical arrow on each record. The

maximum recorded coastal wave height

t was 21 cm at Yakutat, while the maxima

j recorded at BPR stations were 2.8 cm at

WC9, 2.4 cm at AK7 and 1.6 cm at AK8

.

At AKIO on the slope of the Aleutian

l' Trench, at least one cycle of a

I coherent, relatively low frequency wave

j! of amplitude 0 . 4 cm is seen to arrive

1 at about the tsunami ETA. The observed

! period of 5.2 minutes is also long

; compared to that of the background

energy but consistent with tsunami

periods measured at the other three

stations. Tsunami detection at this

station is therefore possible, but

nonetheless uncertain, since the

observed amplitude of 0.4 cm is of the same order as the background noise

level

.

The records at AK7, AK8 and WC9 display

tsunami wave systems which are classically dispersive and amplitudemodulated to form distinct wave

packets, as predicted by theory (see,

e.g., Kajiura, 1963, or Mei, 1983)

.

Tsunami first motion is negative at AK7

and AK8, and this is consistent with

subsidence in the southwest quadrant of

the computed seafloor displacement

pattern. The SNR at WC9 is too low to

be certain of tsunami first motion, but

it does appear to be negative;

interpreting this observation in terms

of the model seafloor displacement is

difficult in this case, since the

source/station geometry is such that

refraction effects may be important.

Finally, it is also of considerable

interest to note the apparent detection

of seismic surface wave energy in the

BPR records, a few minutes after the

earthquake main shock (Figure 2) . Wave

velocities estimated from the time of

arrival at stations AK7 and AK8 were 3-

4 km/s, in good agreement with speeds

which might be expected for Rayleigh

(Rl) waves and/or vertical shear (SVl) waves in oceanic crust (Brune, 1979) . The data are clearly aliased, but

deployment of the new generation of

BPRs characterized by a 15 s sampling

interval promises to provide improved

resolution of these low-frequency

seismic phenomena. >

. ,

5 . CONCLUS I ONS ...

Theoretical and numerical capabilities

in tsunami research have outgrown the

existing observational data base, which

is composed primarily of coastal tide

gauge records that suffer from

nonlinear distortions, inappropriate

time and height scales, the frequent

presence of energetic local resonance

phenomena, and low SNR for small

amplitude tsunamis. High quality

tsunami measurements are now needed in

the deep ocean and at exposed coastal

locations if our understanding of

tsunami generation, propagation and

inundation are to be improved. Small

tsunamis are statistically common in

the Pacific basin, and their accurate

measurement would be extremely valuable

55
to the tsunami research effort . PacTOP

has successfully acquired excellent

deep ocean tsunami records

characterized by maximum amplitudes of

only a few centimeters, clearly

demonstrating the feasibility of longterm deep ocean tsunami monitoring with

self-contained bottom pressure recorder

units

.

6 . ACKNOWLEDGEMENTS

U.S. coastal tide gauge records were provided by L. Hickman and S. Gill of

NOAA's National Ocean Service Tidal

Analysis Section, and the Langara tide

gauge record was provided by F.

Stephenson of the Tides and Currents

Section, Canadian Hydrographic Service.

Tsunami travel time estimates were provided by W. Mass and R. Silcox of

NOAA's Pacific Tsunami Warning Center.

This work is contribution No. 1195 of

the Pacific Marine Environmental

Laboratory's Tsunami Project.

7 . REFERENCES

1. Bernard, E.N. and R. Goulet, 1981:

Tsunami Research Opportunities.

National Science Foundation,

Washington, D.C., 50 pp.

2. Brune, J.N., 1979, Surface waves and crustal structure, Geophys

,

Monogr . I No. 13, The Earth's Crust

and Upper Mantle, P.J. Hart (Ed.),

Amer. Geophys. Union., pp. 230-

242 . 3. Eble, M.C., F.I. Gonzalez, D.M.

Mattens and H.B. Milburn, 1989,

Instrumentation, field operations,

and data processing for PMEL deep

ocean bottom pressure

measurements, NOAA Tech. Memo ERL

PMEL-89, 71 pp.

4. Gonzalez, F.I., E.N. Bernard, H.B.

Milburn, D. Castel, J. Thomas,

J.M. Hemsley, 1987, The Pacific

Tsunami Observation Program

(PacTOP), Proc. 1987 Inter.

Tsunami Symp., lUGG , pp. 3-19.

5. Gonzalez, F.I., M.C. Eble, E.N.

Bernard, H.B. Milburn and D.M.

Mattens, 1990a, Deep Ocean and

Coastal Tsunami Observations: The

1987-88 Alaskan Bight Earthquakes,

submitted to J. Geophys. Res.

6. Gonzalez, F.I., C.L. Mader, M.C.

Eble and E.N. Bernard, 1990b, The

1987-88 Alaskan Bight Tsunamis:

Deep Ocean Data and Model

Comparisons, submitted to J. Natural Hazards.

7. Kajiura, K., 1963, The leading

wave of a tsunami. Bull.

Earthquake Res. Inst., University

of Tokyo, 41, pp. 525-571.

8. Lahr, J.C., R.A. Page, CD.

Stephens and D.H. Christensen,

1988, Unusual earthquakes in the

Gulf of Alaska and fragmentation

of the Pacific plate, Geophys

.

Res. Lett., 15, pp. 1483-1486.

9. Lisowski, M. and J.C. Savage,

1989, Deformation in the Yakataga

Seismic Gap, Southern Alaska,

associated with the Gulf of Alaska

Earthquakes of November 1987 and

March 1988, (Abstract) EOS

Transactions, 44, 1439.

10. Lockridge, P. A., 1987: Historical

Tsunamis in the Pacific Basin. In

Natural and Man-Made Hazards : Proceedings of the International

Symposium held at Rimouski and

Quebec City, Canada, 3-9 August,

1986. M.I. El-Sabh and T.S. Murty

(eds.), D. Reidel Publishing Co.,

pp. 171-181.

11. Mei, C.C., 1983, The Applied

Dynamics of Ocean Surface Waves,

Chapter 2, John Wiley and Sons,

New York, .740 pp.

12. Nishenko, S.P. and K.H. Jacob,

1990, Seismic potential of the

Queen Charlotte-Alaska-Aleutian

Seismic Zone, J. Geophys. Res.,

95, pp. 2511-2532.

13. Okada, Y., 1985, Surface

deformation due to shear and

tensile faults in a half-space.

Bull. Seism. Soc. Am., 75, pp.

1135-1154

.

14. Savage, J.C. and M. Lisowski,

1986, Strain accumulation in the

Yakataga seismic gap, southern

Alaska, J. Geophys. Res., 91, pp.

9495-9506.

56
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58

Ik
Measurement and Modeling of Coastal Storm Effects

BY

1 2 3 William E. Roper , H. Lee Butler , and Andrew Garcia

ABSTRACT

This paper describes techniques used

by the Coastal Engineering Research

Center (CERC) , U.S. Army Engineer

Waterways Experiment Station, for

measurement and modeling of coastal

storm effects. Use of timedependent numerical models for

simulating water levels and flows in

open coastal waters, estuaries,

bays, or lakes has become a standard

practice. This practice has been

extended to modeling wave conditions

generated by the storm and beach

response to storm attack.

Historical CERC research on model

development and measurement advances

is discussed by relating these

advances to practical investigations

performed for various Corps

Districts and Divisions. In

addition, the paper presents a synopsis of the recent storm,

Hurricane Hugo, which devastated the

coastline and infrastructure of

South Carolina. Research teams were dispatched to the scene to

obtain high quality measurements of

high watermarks, beach profiles, and

structural damage.

KEYWORDS: Storm Surge; Waves;

Hurricane; Numerical Modeling; Beach

Response; Erosion; Measurement

Techniques; Hugo.

1.0 INTRODUCTION

The U.S. Army Corps of Engineers, as the U.S. Government agency responsible for protection of the coasts of

the nation, must be able to accu- rately predict coastal flooding from

storms in order to make sound

engineering decisions regarding the

design, operation, and maintenance

of various coastal projects. This

mission requires the acquisition of

high quality field data for the

validation of predictive numerical

models of surge, waves, and beach

erosion. ^

Efficiency in design of coastal

protection is becoming more and more

important. Development of coastal

regions, costs of damages from

storm- induced water levels, and

costs of protection from these

waters are all increasing. Adequate

protection for coastal regions is

desired; however, due to monetary

constraints, the amount of water

level protection that can be con- sidered adequate becomes a question

for which there is no easy answer. Therefore, inherent in any coastal

protection project (new or rehabilitation effort) , there is a need to

develop the best possible estimate

of stage- and eros ion- frequency

relationships for the project area,

as well as an estimate of the error

in these relationships. Research

and development in the area of water

level and beach erosion prediction

is directly linked to the needs of

the Corps District user who has the

responsibility of providing economical and practical designs for each

coastal project.

This paper briefly presents the type

of models and approach used by the

Corps in simulating storm events and

their effects as well as on the

advances being made in instrumentation development and field data

Assistant Director, Directorate of

Research and Development, Civil

Works Program, Headquarters, U.S.

Army Corps of Engineers, Washington,

D.C.

2 Chief, Research Division, Coastal

Engineering Research Center, U.S.

Army Engineer Waterways Experiment

Station, Vicksburg, MS.

3 Research Oceanographer , Prototype

Measurement and Analysis Branch,

Engineering Development Division,

Coastal Engineering Research Center,

U.S. Army Engineer Waterways Experiment Station, Vicksburg, MS.

59
collection methods which permit

acquisition of quality data sets for use in model validation. In

addition, a discussion of the recent

storm, Hurricane Hugo, which devastated the coastline of South

Carolina at the end of the summer of

1989, and its impacts is presented.

Also presented are descriptions of

field data collection and modeling

efforts associated with the event.

2.0 FIELD MEASUREMENTS

Recognizing that development and

improvement of storm surge and

nearshore wave hindcast numerical

models are crucially dependent upon

appropriate atmospheric and hydro

dynamic data, efforts within CERC

are focused on acquiring the needed

data. Because these required data

are typically perishable, that is,

of limited or little value if not

acquired during or shortly after a storm, instrumentation and the

logistics necessary to conduct poststorm surveys have to be in place

prior to the time of storm landfall.

During the last decade, CERC has

maintained a nucleus of personnel

experienced in conducting post-storm

surveys. Because the local infrastructure is typically disrupted

after passage of a storm, experience

in conducting post-storm surveys

includes the ability to communicate,

travel, and sustain operations

unassisted in the area for at least

several days. Moreover, the team

must make efficient use of the

limited time available to complete

as many meaningful measurements as possible. This involves rapidly

identifying and selecting accessible

areas that have experienced significant storm effects.

In addition to post-storm high water

mark and beach profile surveys

,

valuable data acquired during an event include wind, tide, and wave

data. Acquisition of wind data is

officially the responsibility of the

National Weather Service (NWS)

.

However, anemometers used primarily

for providing environmental data for

other purposes, such as tracking

exhaust plumes from industrial and

public utility plants, often provide

valuable supplementary information.

For the purpose of verification of

storm curge models, tide gage data

acquired along open coasts have

broader application as the data are less affected by site specific

characteristics. To acquire these

data, the Corps is cooperating with

the National Ocean Service to

establish a coastal tide gage

network designed to withstand

hurricane effects . There ar^

presently ten sites within tuis

network with new sites being

established at the rate of one or two annually.

Acquisition of wave data during

storm events relies primarily upon

having instruments in place prior to

passage of a storm. The Corps

presently acquires wave data at

approximately 50 nearshore sites

along the oceanic coasts of the

continental United States and within

the Great Lakes. Data acquired by

NOAA buoys located farther offshore

supplement those acquired by the

Corps. The remote sensing of

surface waves , such as performed by

satellites, may at some future time

complement the existing wave data

acquisition network. There also is

the possibility of developing a wave buoy which may be deployed in the

path of an approaching storm from an aircraft

.

3.0 NUMERICAL MODELING

In coastal regions, storm surge

analysis is essential to proper

planning and design of engineering

works and in assessing the extent

and levels of flooding. The theoretical approach to estimating the

frequency and levels of flooding in

coastal waters has been to apply

numerical models for simulating the

storm- induced water motions.

Earlier studies involved investigating project impacts for a Standard Project Hurricane (SPH)

.

Such a storm is usually defined as a hurricane having a severe combination of values of meteorological

60
parameters (such as the storm

radius, forward speed, pressure

deficit, and track) which give high

sustained wind speeds considered

reasonable for a given location.

Later studies have shifted to using

the joint probability method (JPM)

to establish the frequency of flood

level occurrence.

The typical storm surge model solves

the vertically integrated, timedependent, shallow water wave equations of fluid motion. Such

models as the WES Implicit Flooding

Model (WIFM) developed by Butler (1)

have been employed in several Corps

studies in the past two decades.

These studies included investigations in support of a hurricane

barrier protection plan for Lake

Pontchartrain , a northern boundary

for the city of New Orleans

,

Louisiana (2)

.

In the early 1980' s, a study was conducted to investigate the

frequency of storm plus tide flood

levels along the coast and within

the bays of southern Long Island,

New York (3) . The study involved

the development of two WIFM models,

collection of field data for tidal

calibration, computation of coastal

and back-bay stage frequencies, and

analysis of alternative barrier

(dune) and inlet gate configurations. The JPM approach was used to

determine stage frequencies for

hurricanes, extra- tropical storms

(northeasters) , and tide events

along the south shore of Long

Island. New statistical procedures

were used to minimize the computational effort while providing

accurate error estimates (4,5).

Following the Long Island study, a request to apply similar technology

to an area just north of Boston,

Massachusetts, was made by the New

England Division of the Corps. The

purpose of the study (5,6) was to

determine flood levels in the backbay reaches of the Saugus- Pines

River system as well as develop the

frequency of water volumes caused by

storm waves overtopping seawalls

protecting a residential area called

Roughans Point. Flooding caused by wave overtopping is an annual

occurrence. In order to determine

the economic viability of possible

protective measures, it was necessary to establish the frequency of

the flood levels caused by wave overtopping for both the existing

conditions and the proposed protective measures.

Wave overtopping is a complex

phenomenon which is governed by a number of sometimes interdependent

variables such as tide, storm surge, wave height and period, and structure height and slope. To study

this problem five models were used

in conjunction. The WIFM model was used to calculate the still-water

level, and a spectral wave model

estimated the characteristics of waves which would attack the structures at the water levels predicted

by the surge model. A physical

model produced an analytical relationship for determining overtopping

rates given water level, wave height

and period, and structure characteristics. A flood routing model

calculated the maximum stage in the

interior of Roughans Point caused by

each simulated event. A probability

model chose events to simulate,

assigned probabilities to those

events and constructed the stagefrequency curves. These investigations resulted in defining the

most economical seawall, berm/

revetment plan which would provide

sufficient protection for events

with a given frequency of occurrence.

Recent model studies have included

investigations for portions of the

coastlines of New Jersey (7) and

California (8) to assist the

evaluation and implementation of

comprehensive shore protection plans

for these coastlines. These studies

included four broad areas: (1)

deepwater wave climate analysis and

nearshore wave refraction, (2)

numerical modeling of long-term

shoreline change, (3) numerical "

; modeling of storm- induced beach

erosion, and (4) development of

stage - frequency relations for the

61
back-bay and open coast and related

frequency of erosion. These studies

represent an integrated attempt to

quantitatively evaluate long- and

short-term coastal processes on a regional scale for use in

engineering design.

The numerical modeling system

currently used by the Corps for

simulating storm processes and

related effects is the Coastal

Modeling System or CMS (9). The CMS was developed by CERC and comprises

a number of numerical models used in

coastal studies. The system provides a means to organize existing

and new CERC modeling software into

consistent, standardized versions,

using established implementation

procedures and data formats. This

in turn permits easier Corps -wide access to the technology. Resident

in system (or planned) are codes for

modeling long wave hydrodynamics

(such as tides, storm surge,

seiches, and tsunamis), tropical

(hurricane) and extra- tropical

windfields , spectral wave models, wave transformation models, and

sediment transport models (longshore

transport, dune erosion, beach mor- phology, and basic sediment

transport)

.

Current research and development is

aimed at improving wind estimation

models and treating the combined

effect of wind generated waves and

surge in a more rigorous manner. Research in several programs as well

as knowledge gained in specific

mission support studies are contributing to the advancement of storm

surge modeling and related effects.

This research includes a major

thrust using both a morphological

approach and the use of a system of

deterministic models for tide,

waves, wave -induced currents, and

sediment transport. Recent advancements in three-dimensional model

development will provide a means of

simulating the vertical structure of

the storm flow field and conse- quently a better estimate of

associated sediment transport.

4 . 1 Synopsis

Hurricane Hugo is the costliest

hurricane in U.S. history. The

American Insurance Association

reports $3.0 billion dollars as a preliminary estimate of insured

property damage for the U.S.

mainland and $1.9 billion dollars

for Puerto Rico and the U.S. Virgin

Islands. Both figures are subject

to upward revision. Hurricane

damage estimates for past storms

often have been two to three times

the insured property damage, thus

the damage total due to Hugo may

exceed $10 billion dollars (10)

.

The first aircraft reconnaissance of

Hugo was made on 15 September 1989

when the storm was located several

hundred kilometers east of the

Leeward Islands. The aircraft

reported a maximum wind speed of 165

knots, the highest winds observed

during the duration of the storm.

Hugo moved on a westerly course that

took the eye across the Islands of

Guadeloupe, St. Croix, and Vieques,

Puerto Rico. After leaving Puerto

Rico, the storm took a more northerly track toward the U.S.

mainland. At the same time the

forward speed of the storm increased

significantly. Because the track at

this time was nearly parallel to the

South Atlantic coastline of the

United States, small changes in the

storm's direction resulted in large

changes in the predicted landfall

site . Thirty-six hours before landfall in

South Carolina, NWS predictions had

the storm coming ashore near the

northern border of the State of

Florida. Just 12 hours prior to

landfall, the eye of Hugo was expected to cross the coastline near

the city of Savannah, Georgia, and

evacuation was undertaken. Late on

the evening of 21 September, Hugo

came ashore at Charleston, South

Carolina, tracked across the state

and turned northward and moved

through the western edge of

Virginia

.

4.0 HURRICANE HUGO Hugo came across the barrier islands

62
fronting the mainland, causing

devastation from just south of

Charleston (Folly Island) to the

North Carolina border, a distance of

about 200 km. In addition to the

coastal damage, Hugo caused substantial tree blowdown across large areas in both South and North

Carolina. The storm generated

numerous small tornadoes which

touched down and cut off trees at a

3 m height across the two states.

Infrastructure damage behind the

beach was severe. Homes located on

the barrier islands were moved from

their foundations and boats were transported across inland bays and

lagoons to the mainland.

Federal and State agencies were mobilized for victim assistance and

damage assessment. In addition,

there was concern for new flooding

from high lunar tides (exceeding 1 m

above normal) expected in midOctober. The dune line had been

destroyed over 95 km of coastline.

A combination of the Federal

Emergency Management Agency

decision, resource and tasking, and

a request for state assistance, led

to building emergency berms along

the seafront for a total distance of

55 km to protect the infrastructure

from more significant damage caused

by the high lunar tide. The mission

proved a great success , saving

millions of dollars of additional

damage at a cost of about $2.8

million.

4 . 2 Field Measurements

For the purpose of delineating

structural and beach damage as well

as documenting the impacts of Hugo

to assist future research efforts,

CERC sent three teams of engineers

,

scientists, and technicians to

survey high watermarks , beach

profiles, and structural damage.

These data will eventually be very

helpful for validation of Corps

storm surge model methodology and

beach response models. It is

necessary for surge height calculations to be very accurate because

an extra 30 cm of height for a seawall can be a very large cost.

Mindful of the need to acquire

accurate data on the storm's impact,

a team was positioned in a North

Carolina mainland town to move in at

first light and begin making high

water mark measurements . Due to the severe infrastructure damage, logistics were near impossible but a great number of measurements were

- completed.

A second team was dispatched to

Myrtle Beach, South Carolina, to

undertake measurement of beach

profiles. The State had collected

over 75 profile measurements within

3 months prior to the storm.

Profiles change little during the summer in this area, so a field

effort directly after the storm

would result in a high quality data

set depicting the effect Hugo had on

the beach and its dune system.

Again, hurricane impacts on beaches

and dune systems are important to

the Corps because project justification is based on flood and/or

storm damage protection. There is a need to know how high a dune needs

to be, or how wide a beach, in order

to survive a given storm surge and waves . In recent years , CERC has

developed numerical models to investigate dune erosion, but there are very little data to verify the

models . Some survey monuments were

lost in the storm but profile

measurements were completed for

those which survived. Additional

data were obtained from both

infrared and visual aerial photography along the beach and as far as

10 km inland.

A third team surveyed structural

projects and beach damage along the

entire affected area. Included were inspections of seawalls and asso- ciated beach elevation, revetments,

barrier island overwash and

breaching, jetties and bulkheads, . dune systems and breaching, and

other structure integrity. Findings

can be summarized as follows . Where

the large dunes held, there was less

upland damage and conversely, substantial upland structural damage

occurred where dunes were destroyed.

The beaches eroded along the entire

63
study area, and in some cases the

old marsh surface was exposed. Some

sand went offshore, and most of it

back onto the island. Coastal

structures had varying degrees of

damage. Most large ones survived,

providing some degree of protection.

Most of the smaller protective shore

front structures were completely

destroyed with heavy damage to

houses behind them. There was little or no apparent damage to the

landward portion of navigation

structures, such as the jetties at

Murrells Inlet, South Carolina.

4 . 3 Model Analysis

Hugo weakened as it crossed Puerto

Rico and appearances seemed it might

not be much of a storm when it made

landfall on the east coast of the

United States. Then it began moving

erratically and strengthening. Hugo

changed direction several times , and

appeared to be heading for landfall

about 45 km south of Charleston,

until another direction change at

the last moment brought the eye

ashore at Charleston.

Because of the counterclockwise wind

motion in a hurricane, the maximum

winds and largest surge and waves

are on the right side of the eye as

it moves forward toward the shoreline. The maximum winds for Hugo

(shortly before landfall) were about

45 km north of the eye of the storm.

If the eye of the storm had made

landfall 45 km south of Charleston,

instead of passing directly over the

city, the surge at Charleston would

have been much higher.

CERC performed computer simulations

of Hugo using both a wind model and

surge model (NOAA's SLOSH model

(11)). These simulations produced

results which can be graphically

visualized to show the hurricane

leaving Puerto Rico and weakening,

the path meandering, then

strengthening, and finally

collapsing after making landfall.

The models were used to investigate

what would have happened if the eye

of the storm made landfall 45 km

south of Charleston instead of

directly over the city. Hugo

actually produced maximum surge

levels north of Charleston in a relatively unpopulated area. High

water mark measurements gave a maximum surge of over 6 m above mean

sea level. Simulating the same

event, but with a landfall 45 km

south of Charleston, the entire city

would have been under water. The

actual measured surge at Charleston

was 3.2 m above mean sea level. The

simulated surge for a hurricane

landfall south of Charleston gives a surge height greater than 6 m. There would have been hurricane

force winds and waves on top of that

surge

.

Charleston is a relatively small,

low-lying city, with a population of

about 70,000. A 6-7 m surge would

have been more disastrous for a larger city like Jacksonville,

Florida, which has a population of

about 500,000. The last 20 years

have been a relatively quiet period

for hurricanes, and people tend to

forget past hurricanes. There is

real potential for major disaster

from these storms. This potential

inspires continued research efforts

to improve model and measurement

technology for assisting project

design to achieve successful results

when the project is constructed.

Additional studies will be performed

using the data acquired for Hugo.

This event has provided a unique

opportunity to investigate the

accuracy and dependability of models

for simulating wind, surge, wave,

and beach erosion phenomena.

Continuing studies will include

modeling the hurricane windfield for

input to a detailed surge model.

The surge model will simulate

combined tide and storm effects by

using tidal constituent data by

Schwiderski (12) plus an elevation

due to the inverted barometer effect

for open ocean boundary conditions

.

A spectral wave model will be used

to simulate wave conditions

generated by Hugo. Combining these

computations with the surge plus

tide simulation, an estimate of the

total water level produced by the

64
storm can be made. These data will

be archived in a database for later

forcing a storm erosion model which

will be applied at the sites where

profile data were taken. All of the

model simulations will be made as a "blind" test of model accuracy and

dependability

.

5.0 ACKNOWLEDGEMENTS

The research presented in this

paper, unless otherwise noted, was sponsored by the Office of the Chief

of Engineers, U.S. Army, and is

published by its permission.

6.0 REFERENCES

(1) Butler, H.L., "Evolution of a Numerical Model for Simulating LongPeriod Wave Behavior in OceanEstuarine Systems," Estuarine and

Wetland Processes with Emphasis on Modeling . Marine Science Series,

Vol. 11, Plenum Press, New York, NY,

1980.

(2) Butler, H.L. , "Finite

Difference Numerical Model for LongPeriod Wave Behavior: With Emphasis

on Storm Surge Modeling,"

Proceedings of the 1st National U.S.

Army Corps of Engineers - Sponsored

Seminar on Two -Dimensional Flow

Modeling, The Hydrologic Engineering

Center, Davis, CA, 7-9 Jul 1981.

(3) Butler, H.L. and Prater, M.D.,

"Innovative Determination of Nearshore Flood Frequency," Proceedings

of the 20th International Conference

on Coastal Engineering, ASCE,

Taipei, Taiwan, 9-14 Nov 1986.

(4) Prater, Mark D., Hardy, Thomas

A., Butler, H. Lee, and Borgman,

Leon E., "Estimating Error of

Coastal Stage Frequency Curves,"

Proceedings of the 19th International Conference on Coastal

Engineering, ASCE, Houston, TX, 3-7

Sep 1984.

(5) Butler, H. Lee and Hardy,

Thomas A. , "Practical Approach for

Determining Stage Frequency,"

Proceedings of the 18th Joint

Meeting, Panel on Wind and Seismic

Effects, UJNR, Gaithersburg , MD , 12-

15 May 1986.

(6) Hardy, T.A. and Crawford, P.L.,

"Frequency of Coastal Flooding at

Roughans Point, Broad Sound, Lynn

Harbor, and the Saugus- Pines River

System," Technical Report CERC-86-8,

U.S. Army Engineer Waterways

Experiment Station, Vicksburg, MS,

1986.

(7) Kraus , Nicholas C., Scheffner,

Norman W., Hanson, Hans, Chou, Lucia

W. , Cialone, Mary A., Smith, Jane

M. , and Hardy, Thomas A. , "Coastal

Processes at Sea Bright to Ocean

Township, New Jersey," Misc. Paper

CERC-88-12, U.S. Army Engineer

Waterways Experiment Station,

Vicksburg, MS, 1988.

(8) , "Coast of California

Storm and Tidal Wave Study, State of

the Coast Report, San Diego Region,"

in publication, U.S. Army Engineer

District, Los Angeles, Los Angeles,

CA.

(9) Cialone, Mary A., "Coastal

Modeling System, Theory and

Documentation," Instructional

Notebook in publication, U.S. Army

Engineer Waterways Experiment

Station, Vicksburg, MS.

(10) Lawrence, Miles, "Preliminary

Report Hurricane Hugo 10-22

September 1989," National Hurricane

Center, NOAA.

(11) Jelesnianski , C.P., Chen J.,

Shaffer, W.A., and Gilad, A.J.,

"SLOSH-A Hurricane Storm Surge

Forcasting Model," Oceans '84 , Inst,

of Electrical & Electronic

Engineers, New York, NY, 1984.

(12) Schwiderski, E.W. and Szeto,

L.T., "The NSWC Global Ocean Tide

Data Tape (GOTD) , Its Features and

Application, Random-Point Tide

Program," Technical Report NSWC TR

81-254, Naval Surface Weapons

Center, Dahlgren, VA, 1981.

65
Storm Surge Modeling in the United States

by

Chester P. Jelesnianski , Jye Chen and Wilson A. Shaffer

Techniques Development Laboratory

Office of Systems Development

National Weather Service, NOAA

Silver Spring, MD 20910

ABSTRACT

The National Weather Service of the U. S. has

developed the ^ea, l.ake, and ^verland purges

from Jlurricanes, (SLOSH), model to forecast

storm surges whenever a tropical storm threatens. The model is applied to 33 basins. Fig 1,

along the Gulf of Mexico and the Atlantic

Coasts of the United States. It is being

applied to sections of the People's Republic of

China, India and the islands of the Bahamas.

The model is also used as a tool for storm

evacuation planners to delineate areas of potential flooding.

1. INTRODUCTION

SLOSU—is a numerical-dynamical computer model

designed to compute tropical storm surges. The

model was designed in the National Weather Service's Techniques Development Laboratory for

"real-time" forecasting. The storm surge—

a

significant increase in water level accompanying a storm—has long been known as the major

threat to life. In one incident alone, storm

surge from the 1900 Galveston hurricane

resulted in 5000 to 6000 deaths.

Tide gage observations taken during a storm's

passage show that the storm surge lasts, typically , about 6 hours and affects about 100

miles of coastline. Very large, slow-moving

storms can produce significant surges for a much longer period and affect a longer stretch

of coastline. To emphasize the threat posed by

the surge, consider the surge from Hurricane

Camille in 1969 which reached 24 feet, or the

surge from Hurricane Hugo in 1989 which reached

20 feet.

2. THE SLOSH MODEL

The SLOSU model is two-dimensional, covering a segment of the continental shelf, inland water

bodies and terrain. The equations of fluid

motion are solved numerically, incorporating

finite amplitude effects but not the advective

terms from the equations of motion. SLOSH uses

a time-history bottom stress (Platzman, 1963;

Jelesnianski, 1967), corrected for finite am- plitude effects. At any given point, the computed surge is designed to reproduce the timehistory of a long-period gravity wave as shown on a tide gage hydrograph or stage record.

Short period phenomena, such as wind waves and

their associated "run—up", are ignored.

Most SLOSH basins use a polar grid, such as the one shown for the Lake Pontchartrain/New Orleans basin Fig. 2a. This polar grid allows a fine mesh in the primary area of interest. One

advantage of this grid is that boundary conditions are imposed far from the area of interest. Other basins use an elliptical/hyperbolic

grid which have special properties that can be

used advantageously. Figs 3(a,b).

On every grid "square" of the model, a value of

terrain height or water depth is supplied, thus

tailoring the model to a specific basin area. Literally hundreds of maps (USGS quadrangle or topographic maps, NOS bathymetric charts, and

various types of plats) are used to specify

these values. The model incorporates barriers

which impede the flow of water and allows for

the overtopping of these barriers. Such barriers, include dunes, levees, spoil areas, natural ridges, reefs, and various man-made structures. The model also treats sub-grid size

flow through cuts between barriers, channel

flow, and width variations along rivers.

As the computations proceed in time and if water overtops barriers, water spills into areas

behind the barriers and is acted on by storm

driving forces. Model computations are turned

on for newly wetted grid squares. Likewise, as water recedes, computations are turned off on dried grid squares. Water penetrates inland

until impeded by other barriers or naturally

rising terrain. It is possible for a storm to produce massive inundation across low lying

terrain, extending many miles inland.

Imbedded within SLOSH is a tropical storm wind

model. For a computational run of SLOSH for a given storm event, the user must supply simple,

time-dependent meteorological parameters.

66
These are: position (latitude and longitude)

of the hurricane, central pressure, and storm

size (distance from storm center to the maximum

wind)

.

These inputs are entered at 6-hour intervals,

beginning 48 hours before landfall and ending

24 hours after landfall. In the event the

storm does not make landfall, the time of the

storm nearest to a basin's defined origin is

used instead of landfall time. Note that wind

is not an input parameter. The SLOSH storm

model produces a vector wind field throughout

the basin by balancing forces according to the

meteorological input parameters. Fig. 4. depicts a computed surge envelope of highest

waters generated by hurricane Betsy, 1965, for

the Lake Pointchartrain basin. The surge contours were drawn from output surge data by a line printer on the image or computational

plane. Fig. 2b. This is a typical result when

computing with historical storms for any of the

basins of Fig. 1.

3. MODEL ACCURACY

We attempt to simulate inland flooding with

historical storms in order to verify the SLOSH

model. Surge data consists of tide gage observations, staff gage records and high water

marks. High water marks inside buildings or structures are the least accurate due to con- tamination by wave action and inadequate damping or overdamping of the water levels by the

structure. Often high water marks vary by 20%

for two nearby locations.

Before a SLOSH simulation, the hurricane's

track is determined as precisely as the data

allow. In addition to the track, we estimate,

the storm's radius of maximum wind and its

central pressure.

In Fig. 5, we show the combined results of

simulations for several basins. The error is

generally within + 20% for the significant

surges, with a few observations falling outside

that range. A total of 572 surge observations

were used to develop this figure. These observations were taken throughout the area affected

by the surge—around the maximum surge, on inland terrain, at the storm's periphery, and

along inland water bodies. Note that the tide

gage data are limited to the lower observed

heights; tide gages frequently fail during

major surge events.

The SLOSH model is not "tuned" for a particular

geographic location, but uses "universal" specifications for constants such as the model's

drag coefficient, bottom stress, etc. This

allows us to adapt SLOSH to any geographical

location, whether or not it has ever experienced a storm, and have confidence in the model's computations.

4. EVACUATION PLANNING WITH SLOSH

Although SLOSH originated as a forecast model,

it is also used as a tool to delineate areas of

potential storm flooding. Witli this information, an evacuation planner can identify areas

for evacuation, determine which highways can be

used for evacuation routes, and site shelters

in safe areas.

To find a region's potential for storm surge

flooding, a large number of hypothetical storms

are simulated to impact the area, with the

flooding due to each one archived. These

storms are varied in intensity, size, and landfall point along climatoiogically likely

tracks. Typically, several hundred storms are simulated in a basin. This number is sufficient to highlight critical storm paths that may

pose excessive flooding in an area.

A spinoff of the simulations is an "atlas" of

flooding within a basin. This atlas comprises

displays of flooding for each storm run for Llie coast, a forecaster can turn to the atlas,

match his forecast storm (path, size, and intensity) to one already in the atlas, and get

an approximation of the flooding. due to that

s torm.

The atlas also contains composites of flooding.

For example, areas flooded by a storm coming

from a given direction, regardless of its landfall point, and having a given intensity, may

be combined. This composite aids the forecaster by pointing out critical regions where flooding may be extreme.

The immense data base for storm evacuation

planning are stored in Micro PC's for instant

recall in colorized, graphical output. This is

a continuing development in the National Weather Service of the United States.

4. REFERENCES

1. Jelesnianski , C. P., 1967: Numerical com- putations of storms surges with bottom

stress. Mon. Wea. Rev

.

, 95, 740-756.

2. Platzman, George W. , 1963: The dy[iaraical

prediction of wind tides on Lake Erie.

Meteorological Monograplis , Amer. Met. Soc .

, 4, 44 pp.

67
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74
Theme III

Joint Cooperative

Research Program
Ii i ;\

'I

n'l I

'l I
The Japan Research Plan on Precast Concrete Seismic

Structural System (PRESSS Japan)

by

ShinOKAMOTO*

Shinsuke NAKATA*

HisahiroHIRAISHI*

Takashi KAMINOSONO*

Hitoshi SfflOHARA*

* : PRESSS Project Team

Building Research Institute

Ministry of Construction

ABSTRACT

This report is the research project plan from Japanese side on Precast Concrete

Seismic Structural System (PRESSS) for U.S. - Japan Cooperative Research Program

which started since 1981 . This PRESSS program is the fourth phase item under the auspice

of the UJNRwhich will continue for four years from this fiscal year .Our Japanese side

program include not only the academic research development but also the establishment of

design guideline and the design manual for precast concrete joints . Through this project , a

feasibility study on "hinge-device system" is also discussed.

77
1. Background & Objectives

Precast concrete buildings have been widely constructed since 1960's in Japan. In 1965,

Architectural Institute of Japan published a design recommendation for wall-panel precast

concrete building. Since then, the number of precast wall panel concrete building increased

responding to the increasing demand of urban housing supply. In 1974, the Japan Building

Center established "Guideline for High-rise Precast Wall Panel Reinforced Concrete

Structures", the scope of which is extended up to 8 storied building and it is based on

ultimate strength concept. Most recently, aiming at the savings of shore and reduce

construction period, half-precast construction systems, which use prefabricated beam and

slab and made monolithic by topping concrete at the construction site, has been developed

and widely used now.

In addition to that, the problem of shortage in numbers of construction labors is urgent in

Japan. Construction companies try to rationalize and reduce the need of number of labor at construction site. Prefabrication is one of the solution of the problem. Some construction

companies have already developed precast frame building system. However the systems are

different each other. Thus, construction plan using the precast building systems require

special screening by the Japan Building Center at present. It cause a delay of construction

schedule and prevent construction of precast building from extensive usage. Furthermore,

lack of detailed design recommendation is an another problem.

Taking into consideration of these Japanese conditions, the development of commonly

used design method of precast concrete frame building is indispensable. U.S. -Japan

Coordinated Research Program on Precast Concrete Building will cover new guidelines for

design and construction. The coordination between United States and Japan will make it

possible to develop new guideline for precast concrete buildings more earlier and more

effectively

.

2. Scope of Research

The scope of this research project is as follows:

(1) Type of Structural System

Moment resistant frame system and Moment resistant frame with shear wall system

will be investigated

(2) Limitation of Building Height

Building with less than 15 stories ( less than 45 meters )

78
(3) Material Strength

Concrete Design Strength fc< 500 kg&'cm^

Longitudinal Reinforcement Yield Strength ft 3000 - 5000 kgl/cm^

Lateral Reinforcement ft 3000 - 13000 kgf7cm2

(4) Details of Connection of Precast Member

In Japan so many types of connection are used in practice varied that it is impossible

to test all of existing connection systems within the project. Therefore, all of

cormection systems are not necessary included in the scope of this project. The

selection of connection systems will be done according to mechanical characteristics of

each connection systems.

(5) Nonstructural walls ' .

Nonstructural walls or cladding of buildings are dealt with only from the viewpoint

that they should not have an unfavorable influence to the structural members.

3. Target of Research Development

The objective of this project is to develop a methodology of earthquake resisting design

and a methodology for quality control of precast concrete building during production and

construction.

The target for development is moment resisting frame building made of precast concrete

which have high flexibility for planning design, as much as monolithically constructed

reinforced concrete buildings.

The Philosophy of seismic design will be based on design philosophy of ductile moment

resisting frame. Design philosophy may be followed from recently revised draft code named

"Design Guideline for Earthquake resisting Reinforced Concrete Building Based on Ultimate

Strength Concept," published from Architectural Institute of Japan in 1988.

For precast concrete buildings, several different design philosophy exists which include :

(1) CAST-IN-PLACE EQUIVALENT DESIGN

(2) SLIP ALLOWING DESIGN

(3) HINGE DEVICE DESIGN

"CAST-IN-PLACE EQUIVALENT DESIGN " is design concept which make possible to design building replaceable by precast construction without significant change of design.

"SLIP ALLOWING DESIGN" is design concept which allows occurring slip behavior in

hysteretic loop due to connection characteristics by compensation of increased strength.

Intensive study of non-linear dynamic response analysis considering slip type hysteresis will

79
needed. "HINGE DEVICE DESIGN" is design concept which expect and utilize hysteretic

energy absorption at specially designed connector.

In this project, design concept of "CAST-IN-PLACE EQUIVALENT DESIGN" will be

focused on and practical guideline based on this philosophy is planed to be developed.

Design (2) and (3) are planed to be studied in future in accordance with mutural coordination

between U.S. and Japan PRESS.

4. Research Products

As a result of research in this project, several design recommendation and manuals are planed to developed, especially concerning with "CAST-IN-PLACE EQUIVALENT

DESIGN". These design recommendation and manuals should cover not only existing joint

construction systems, but also new types of devices flexibly, which will be developed in the

future. Research products are as follows;

( 1) Development of design recommendation for precast concrete building

This design recommendation covers moment resisting frame with or without shear wall to

fifteen storied buildings. The guideline is utilized in practice by structural designers and

buildings superintendent in specific governmental offices where check and evaluate design

procedure before construction of buildings. This recommendation should describes required

seismic performance and the rules for calculation procedure. The contents will be as

following;

1) Scope

2) Structural planning

3) Stiffness evaluation

4) Design loads for members with joint region

5) Required strength and ductility for members and joint regions

(2) Development of Design Manual for Precast Member Connections

Design manual for Precast Member Connection is planed to be developed in order to show design procedure, by which designer will able to design connection with satisfactory

performance with respect to strength and ductility.

This design manual should be practically used among structural designer, makers of

precast , constructors, constructing managers, and inspectors in governmental office.

80
This manual should describe required performance ofjoints, design procedure of detail , and the technical background data useful for joint design as an appendix.

The contents of will be as following;

1) Scope

2) Design loads and design required ductility for joints

3) Fundamental rules of design for members with joints

4) Insurance ofjoint strength

5) Insurance ofjoint ductility

6) Design ensuring durability

7) Production and transportation

8) Site construction

Appendix A: design example of members and joints

Appendix B: Joint data

Appendix C: Evaluation Standard ofjoint performance

(3) Establishment of guideline for site construction and erection

The guideline for production and erection contains the items which make us of precast

concrete members and site constructors should follow so that constructed precast concrete

buildings can keep required quality in the process of their production and erection.

This guideline does not contain the items of a qualification check for makers and

constructors on quality insurance.

5. Research Plan

5.1 Development of Design Recommendation for precast concrete building

( 1) Making Frame Work of Design Recommendation

For the first stage, it is necessary to develop the frame work. Frame work means a logical

structure of design recommendation. By the examination of the frame work, the scope of

research work will be clearly defined and understood. A compilation of existing design

philosophy which is used by precast engineer will be also necessary in this preliminary

stage.

Thus the item which is needed to investigate at first stage will include:

- How to modify current three earthquake resisting design procedure; so called route

1, route 2 and route 3, i.e.

81
route 1: one of alternate design method of route 3, route I does not require the

check of lateral strength capacity, which is a kind of strength oriented design. The

scope of route 1 is limited to low-rise building which have enough amount of

lateral resisting member, and minimum ductility is required.

route 2: another alternate design method which does not require the check of

lateral strength capacity, which rely on ductility to some extent.

route 3: all-mighty procedure but which require the check of lateral strength

capacity.

- How to handle buildings which have both monolithic members and precast

members. This problem will arise because the scope of route 1 , route 2 and route

3 are defined separately corresponding to different structural system, i.e. reinforced concrete, steel and so on.

- Examination of existing design philosophy of precast wall panel construction

design, which is currently used by Japanese design engineers

(2) Making Draft of Design Recommendation

Then the details of design recommendation should be made based on analytical and

experimental research. The problem which should be solved will include:

- The scope and limitation of structural system

-The scope and limitation of number of story, eccentricity, distribution of story

stiffness in height direction

- Lateral strength demand for precast structural system

- Strength and ductility demand to joint which connects between precast members

including two types of connection as follows;

i) connections used in building designed by capacity design concept

ii) connections used in building designed by traditional design concept

- The design philosophy of imperfectly integrated precast floor diaphragm

transferring in-plane force induced by earthquake action

- The design philosophy of serviceability

(3) Experimental Studies on Effectiveness of Design Recommendation

Tests of full scale three-story three-frame specimens representing the bottom three stories

of fifteen-story precast frame structure are planned to be designed. The objects of this

experimental research program is summarized as follows;

- Correlation study between analytical model and experimental result

82
- Check of construction manual for quality control , which is planed to be

developed as the fruit of cooperative research program between BRI and

member from precast industry.

- The check of the design manual for precast member connection which is also

planed to be developed in this project.

- Total Verification of Precast structural system.

5.2 Development of Design Manual for Precast Member Connections

In Japan, no design manuals for precast member connection exists, which describes the

philosophy as well as appropriate example of many kind of connection details. So we need

such manual which can be helpful for precast concrete engineer.

The manual will describe a design methodology of connection which transfer stresses

safely and have enough ductility to cyclic loadings under the combination of design load.

Experimental and analytical research will be done for the purpose of obtaining basic data

for documentary of design manual for precast member connection. The experimental research

will be divided into three categories as follows;

(1) Tests of Sub-element Consisting Precast Connection

Joint materials and interfaces are defined here as sub-element of precast connection. The

stress condition of sub-element are rather simple than that of whole connections. Thus the

data about the sub-element can be basic in design of connections.

The sub-element should be identified and classified. Then the sub-element will be tested

under monotonically increasing load. Temporally idea for tests of the sub-element will

varied as follow;

Sub-element in Beam and column connections:

- concrete interface test considering the stress at the connection of beam to column

or beam to beam subjected to shear as well as bending moment axial load,

(variables: roughness of concrete interface, amount of dowel bars, inclination of

compressive strut which cross the discontinuous interface, thickness of leveling

mortar, axial load level, strength of leveling mortar, etc.)

- bond test of longitudinal reinforcement in beam which lay along the interface

between precast concrete and topping concrete. (variables: the shape of cap tie,

diameter of longitudinal bars, amount of cap tie, etc.)

- mechanical integrity of T-shape section beam consisted of precast beam and in-situ

topping concrete (variables; amount of reinforcement across concrete interface)

Sub-element in Shear Wall:

83
- interface between wall panel and boundary element (variables: inclination of

compressive strut to interface, amount of shear reinforcement, roughness of

concrete interface, shape of shear cotter, etc.)

(2) Test of Precast Member Subassembledges

Based on the test data of connection materials and interface, analytical model will be

developed which predict strength of connections. In order to assure the correlation between

prediction of model and test, test of subassembledges of precast member is needed. For

these subassebledge tests, cyclically loading which simulate earthquake will be used in order

to evaluate ductility of connection.

Through these tests, the concept of connection design will be proposed and verified.

Subassembledge tests will include:

- column:

the effect of axial load, shape of cotter, dowels to ductility will be examined

- beam:

concrete interface at beam end

the method of consideration of dowel effect

concrete interface between precast concrete and topping concrete

shape of cap tie, amount of cap tie

- shear wall

vertical joint

horizontal joint

joint wall panel and beam

- beam-column-joint subassembledge:

anchorage method of beam longitudinal reinforcement

- floor system

connection of slab panel

- connection of beam to girder

beam end reinforcement in stress discontinuous zone

(3) Test of Superassmbledge

One-bay one-story frame specimens will be tested. The object of this test is as follows:

- Experimental verification of design of collapse mechanism

- Evaluation of story stiffness

- Correlation study between Model and experiment

84
5.3 Development of Construction Manual for Quality Control

This manual will be developed as the fruit of Cooperative research program between BRI

and member from Precast industry.

5.4 Conceptual Flow of Research

Figure 1 shows the Time schedule and conceptual flow chart of research concerning

development of 1) design recommendation and 2) design manual of precast member

connection.

6. Research Budget and Executive Structure

Building Research Institute, Ministry of Construction of Japanese Government will spend

about 120 Million Yen as a national project in total from 1989 to 1993( in fiscal year).

Parallel to this project, JTCC-PRESSS have been negotiating with Japanese

constructors and precast concrete industry to carry a cooperative research. If it is possible,

they will expense necessary for additional research and which will be able to support this

project separately.

Figure 2 shows tentative research executive structure of Japanese side. It has two main

committee, the functions of which is as follows;

(1) Advisory Panel Promoting Cooperative Research

a. approval of cooperative research executive structure

b. approval of cooperative research plan

c. approval of policy of compilation of research result

d. approval of policy concerning publishing of research result and

treatment of patent.

e. promoting and utilization of the accompHshed results of Project

(2) Joint Technical Consulting Committee (JTCC-PRESSS)

a. coordination of research and development targets with U.S.-PRESS

committee

b. coordination of research plan of Japan side

c. coordination of utilization policy of research result with U.S-PRESS

committee

d. advice researches on carrying out of research works

e. critical review of research result

85
7. Research Schedule

The Project of BRI financially supported by Ministry of Construction will last for four

years, Thus it will continue since April 1989 until March in 1993. Although Project U.S.

is supposed to become a six-year project starting in 1990. So JTCC-PRESSS should be

extended for more two years. As a counter part of U.S.-PRESSS Committee.

86
989

990

Arrangement of Data for Connections

Stress Transrer Mechanism

Segmentation

Frame Work of Design Guideline

Classification of Connection System

Modelling

Sub-Element

Stress transfer Mechanism

1991

I

Testification

Pilot Test

Subelement

Subassemblage

PCa Assemblages

Strength And Ductility

I

Definition of Mathematical

Modelling -SubassemblagesTest - beam

column

shear wall

criteria case study

Testification of Structural Performance

Sub assemblage, Super assemblage

Modelling

1992

I

Evaluation of Dynamic

Response Analysis

Evaluation Method of Design Strength

of Connections

I

General

Testification

by

Full-scale Test

Design Manual of Connections Design Guideline of

PCa Frame Building

Fig. 1 Conceptual Flow of Research

87
MOC

BRI

Academy Group

B.CS.

PCa Industry

I HUDC

Conclusion of Agreement

on the Coordinated

Research Program

( UJNR)

NSF

JTCC

PRESSS

once a year

Research Promotion

Panel

once a year

Technical

Coordinating

Committee

( 20 members ) once / two months

Working Group

on

Design Criteria

/ \

Working Group

on

Joint Design

Manual

Japan Building Center

Technical

Coordinating

Committee

U.S. Side

Fig. 2 PRESSS Executive Structure

88
U.S. Coordinated Program for Masonry Building Researcli

Fifth Year Status

James L. ABSTRACT

The U.S. program, which consists of

twenty-eight specific research

tasks, is coordinated with a parallel program in Japan. Both are conducted under the auspices of the

UJNR Panel on Wind and Seismic

Effects. The primary purpose of the

U.S. program is to support the

development of a limit state design

procedure for masonry as well as provide experimental data,

analytical methods, etc. Several

technical accomplishments have been

made in both modeling and

experimental areas. The technology

transfer plan includes dissemination

of research reports, topical reports

and a summary report on the specific

research areas as well as by

presentations and papers. Design

and criteria recommendations will be an extremely important mode of

technology transfer and are expected

to support design standard

development

.

KEYWORDS: Masonry; structures;

standards; experiment; reports.

1 . INTRODUCTION

The U.S. Coordinated Program for

Masonry Building research is an integrated set of 28 specific

research tasks being conducted by

the U.S. Technical Coordinating

Committee for Masonry Research

(TCCMAR/US). The U.S. program is

coordinated with a parallel program

in Japan to exchange information and

concepts for the mutual benefit of

both countries. Both programs are conducted under the auspices of the

UJNR Panel on Wind and Seismic

Effects. The U.S. program is funded

by the National Science Foundation.

2. PROGRAM OBJECTIVES

Primary program objectives are:

a. Development of design and

criteria recommendations for

limit state design of

reinforced masonry buildings

and components.

b. Development of a consistent

experimental database on the

Noland^

behavior of masonry materials,

components and systems.

c. Development of analytical non- linear models for research and

design office use for detailed

analysis, system analysis, and

dynamic loads determination.

d. Improved material and

subassemblage experimental

procedures for obtaining

masonry properties.

e. Improved masonry fabrication

procedures and standards.

f. Developing an increased

awareness among engineers,

architects, code bodies and the

public of the capabilities of

reinforced masonry in all

seismic zones.

g. Interfacing with standards

development groups to support

development of a consensus

limit state standard for

masonry.

3. APPROACH

With NSF support, the Technical

Coordinating Committee for Masonry

Research/U.S. (TCCMAR/U . S . ) was formed in February 1984.

TCCMAR/U. S. was and is comprised of

researchers from academic and

industrial organizations who have

strong backgrounds in research into

the properties and characteristics

of reinforced masonry materials,

structural components and systems,

analytical techniques, structural

dynamics, building codes, and

earthquake engineering. TCCMAR was

not intended to be a closed group;

researchers may be and have been

added as needs develop. Current

TCCMAR researchers are listed in

Table 1.

The initial TCCMAR purposes were 1) to specifically define the research

topics, both experimental and

analytical, necessary to develop a

1. Atkinson-Noland & Associates,

Boulder, Colorado

89
consistent masonry structural

technology for the U.S. and 2) to establish communication with its

Japanese counterpart to enable

Japanese and U.S. programs to be

coordinated for the benefit of both.

Program evaluation and research

needs assessment are continuing

activities of TCCMAR. The original

plan is not a static entity but

rather one which can and has been

modified as work has progressed and

needs either revised or added.

The U.S. research plan is a phased

step-by-step program of separate,

but coordinated research tasks.

Emphasis is being placed upon intratask information exchange and

consultations so the experimental

and analytical efforts are mutually

supportive.

The program includes both

experimental and analytical research

tasks. The experimental and

analytical efforts are conducted in

parallel with each becoming more complex and sophisticated as knowledge and experience accumulate.

The U.S. plan is thus a "building

block" approach.

The U.S. program is being conducted

on a project basis to provide the

task and schedule coordination

required for efficient and orderly

conduct of the program.

4. SCHEDULE

The schedule herein (Figure 1) contains a list of the specific

research tasks which comprise the

U.S. program and their starting and

ending dates. Primarily because of

funding difficulties the schedule

has been revised. The program has

been extended until January 1994,

i.e., one year after the completion

of the full-scale test project.

5. TECHNICAL ACCOMPLISHMENTS

The U.S. program is proceeding

generally according to expectations.

At this point in time, which is

about the 60 percent point,

technical accomplishments have

become apparent. It must be

recognized, however, that the major

technical results are yet to come.

Because the primary objective of the

U.S. program is to provide a basis

for limit state design and an associated generic modeling

capability, the following seem significant among the results

achieved to date:

a. A finite element model (FEM) has

been formulated and implemented

capable of reproducing some aspects of post-peak strength

response of reinforced shear

walls. The material models

have the capability to represent the reduction in peak

compressive strength due to tensile cracking and allow

reorientation of initial cracks

to a final crack state. The

model is capable of reproducing

fully cyclic load-deformation

histories and is operable on

286 and 386-based PCs.

b. Structural component models (SCH)

have been developed and

calibrated using the results of

the one-story wall tests. Tlie models will soon be extended to more complex stress states, and

will be combined for use in

analyzing pierced walls. The

models will be extended to be

able to handle cases involving

unloading and reloading.

c. A non-linear lumped parameter

model (LPM) has been developed

to do time-history analyses of

masonry structures. New non- linear "spring" elements have

been introduced to represent

masonry shear walls in building

system models. The LPM will

include rotational degree of

freedom, and will be tested

against experimental results.

d. Approximately 30 reinforced

shear walls have been tested.

A database has been developed

for the cyclic behavior of

single-story reinforced masonry

shear walls with different inpiane compressive loads and

reinforcement rates. Limit

states have been identified for

application in the design

recommendations to be made.

e. Three different two-story

coupled reinforced masonry

shear wall specimens have been

tested resulting in a database

for the fully-reversed cyclic

behavior of such specimens with

different degrees of coupling.

90
stable hysteretic behavior was observed at overall story

drifts of at least one percent.

Limit states have been

identified for application in

design recommendations. Simple

models predicted collapse loads

within ten percent and complex

models satisfactorily predicted

collapse and deflections.

Force-deflection characteristics

of prestressed precast concrete

plank floor diaphragms have

been established experimentally

both parallel and perpendicular

to plank direction. Limit

states have been identified and

corroborated by observations of

plank behavior in the two-story

coupled shear wall tests.

Data describing the behavior of

other diaphragm systems, e.g.,

timber, steel deck and steel

deck-concrete, has been

collected. A comprehensive

final report presents the data

to define hysteretic models for

the diaphragms and other design

conditions

.

Reinforced concrete and clay

masonry walls responded

elastically when subjected to out-of-plane dynamic motions

typical of design ground motion

in soil type 1 in the highest

U.S. seismic zones. Limit

states for application in

design recommendations have

been identified.

Tests of grouted hollow clay and

concrete unit prisms in

concentric compression

indicated that masonry is a generic material, i.e., that no unusual behavioral traits are present to separate grouted

hollow-unit clay and concrete

masonry. The compressive tests

established stress-strain

curves including the post-peak

descending branch and limit

state values. Tests of prisms

loaded in eccentric compression

established strain gradient

effects and necessary to define

equivalent rectangular stress

values for use in flexural

design.

Tests have been conducted on reinforcement bond-slip and lap

splice behavior using grouted,

reinforced hollow clay and

concrete specimens. Both pushpull and pull-pull loads were

applied. Results have been

used to formulate models and

identify limit states for

application in design

recommendations

.

k. A design criteria was developed

for shear walls that was adopted in the 1988 Uniform

Building Code, Section 2413.

1. A proposed design criteria was developed for Nasonry Frames

for the 1991 UBC. The criteria

is now under review.

m. A new proposed design equation

was developed for masonry shear

walls. This equation was submitted for inclusion in the

1991 UBC and is now under

review.

6 . TECHNOLOGY TRANSFER

The technology developed, data, and

other findings which result from the

research must be made available to the public for study and use in an active manner in order to have the

greatest utilization possible. For

a program such as this which extends

over a substantial period of time,

it is appropriate that the

technology developed, etc., be

documented and released as an on- going activity as well as the

documentation and dissemination

which will occur at the completion

of the work.

The Technology Transfer component

which has been defined for the

program consists of the following:

a. Task Reports - One or more reports have been and will be

prepared for each research task

comprising the coordinated

program. Task reports are complete technical research

reports. Each task report is a "stand-alone" document and

fully describes the subject of

the report. In the case of

reports on experimental

research, all pertinent data is

included in a clear and

understandable form. Copies of

task reports are distributed

initially to about 75 concerned

individuals. Arrangements have

been made to furnish the

91
Earthquake Engineering Research

Library at UC-Berkeley with

copies of all Task Reports

which have been and will be

released. Completed task

reports are listed in Table 2.

Papers - Papers published in

technical journals and

conference proceedings are an important part of technical

transfer because they focus on individual issues and are more widely read than larger

documents. TCCMAR researchers

have and will continue to prepare such papers.

Arrangements have been made to provided complete copies of the

proceedings of all the meetings

of the Joint U.S. -Japan

Technical Coordinating

Committee on Masonry Research

to the Earthquake Engineering

Research Library at UCBerkeley.

Presentations - Presentations on various aspects of the

Coordinated program to professional, code, industry,

and other concerned groups are encouraged and many have and

will continue to be, made.

Typically, no papers are associated with presentations,

but some material may be

distributed

.

Seminars & Workshops - Participation in seminars and

workshops is another form of

Technology Transfer which is

encouraged because of the more in-depth communication which

can occur. Some participation

has occurred and more is

anticipated

.

Topical Reports - Reports will

be prepared on each basic topic

addressed in the Program and

will be a treatise on the

subject. Anticipated topical

reports are listed in Table 3.

Technical Summary Report - The

technical summary report is to be prepared late in the program

will present basic technical

findings and conclusions and

sufficient supporting data and

information to substantiate

them. Topical Reports will be

the primary references, but

92

task reports will be listed if

additional detail is required.

g. Design^, and .^Criteria

Recommendations - Recommendations for reinforced

masonry building design and

criteria will be formulated

documented in a manner suitable

for review and adoption by code

bodies. Recommendations will

also be made for standard tests

which provide masonry material

properties.

7. STANDARDS/CODES

A critical issue in the U.S. is, and

has been the transfer of technol

developed by research into practice.

In the U.S. an important factor in

implementing research into practice

is the incorporation or provision

for new technology by the design

standards.

In the U.S., design standards are developed by non-governmental

professional groups using various

forms of consensus processes.

Hence, the NSF-suppor ted U.S.

Coordinated Program for Masonry

Building Research cannot develop

standards, but will develop

recommendations (Task 10)

.

Discussions and negotiations over

the last year have served to establish mechanisms and processes

for transfer of technology (i.e.

research data and recommendations)

to a consensus design standards

development committee sponsored by

The Masonry Society, the American

Concrete Institute and the American

Society of Civil Engineers. A oneyear pre-standardization effort to collect and document background

information pertinent to limit state

structural design and to develop a draft limit state design standard

for masonry was organized and funded

by industry. The prestandardization work along with the

technology developed by the

coordinated masonry research program

will be utilized by the national

masonry standards committee to produce a limit state design

standard for masonry. The basics of

the process are illustrated by the

figure 2.

The first meeting of the Joint

Committee was on December 11 & 12,

1989 in Denver. A task group was
organized to formulate a single set

of operating procedures. In addition to developing a limit state

design standard, the Joint Committee

will maintain the recently completed

working stress standard, and address

other issues pertinent to masonry

structures design and construction.

The second meeting of the Joint

Committee was held in Toronto in

March 1990. James Colville of the

Department of Civil Engineering,

University of Maryland was inducted

as the chairman of the Joint

Committee

.

93
TABLE 1

TCCMAR RESEARCHERS

Daniel Abrams

University of Illinois

Urbana, IL

Richard Atkinson

Atkinson-Noland & Associates

Boulder, CO

Robert Englekirk

Englekirk & Hart

Los Angeles, CA

Ahmad Hamid

Drexel University

Philadelphia, PA

Gilbert Hegemier

Univ. of CA-San Diego

La Jolla, CA

Richard Klingner

Univ. of Texas-Austin

Austin, TX

James L. Noland

Atkinson-Noland & Associates

Boulder, CO

M.J.N. Priestley

Univ. of CA-San Diego

La Jolla, CA

P.B. Shing

University of Colorado

Boulder, CO

Samy Adham

Agbabian Associates

El Segundo, CA

Russell Brown

Clemson University

Clerason, SC

Robert Ewing

Ewinq & Associates

Rancho Palos Verdes, CA

Gary Hart

Englekirk & Hart/UCLA

Los Angeles, CA

John Kariotis

Kariotis & Associates

South Pasadena, CA

Ronald Mayes

Computech Engin. Services

Berkeley, CA

Max Porter

Iowa State University

Ames, Iowa

Frieder Seible

Univ. of CA-San Diego

La Jolla, CA

Leonard Tulin

University of Colorado

Boulder, CO

94
TABLE 2 - TASK REPORTS

i

Task No. Author/s - Title

1.1-1: Atkinson and Kingsley, Comparison of the

Behavior _pf Clay and Concrete Masonry in

Compression^ September 1985.

1.2(a)-l: Hamid, A., Assis, G., Harris, H., Material

Models . f pr__Grouted Block Masonry

1.2(b)-l: Young, J.M., Brown, R.H., Compressive Stress

Distribution of Grouted Hollow Clay^ Masonry

Under Strain Gradient, May 1988.

2.1-1: Hart, G. and Basharkhah, M., Slender Wall

Structural Engineering Analysis Computer

Program, (Shwall, Version 1.01), September

1987.

2.1-2: Hart, G. and Basharkhah, M., Shear Wall

Structural Engineering Analysis Computer

Program (Shwall, Version 1.01), September

1987.

2.1-3: Nakaki, D. & Hart, G., Uplifting _Response_ of

Structures Subjected to Earthquake Motions,

August 1987.

2.1-4: Hart, G., Sajjad, N., and Basharkhah,

Inelastic _Column Analysis Computer ,P_rogram

(INCAP, Version 1.01), March 1988.

2.1-5: Hong, W.K., Hart, G.C., Englekirk, R.E.,

Force-Deflection Evaluation and Models for

University of ^Colorado Flexural _Walls,

December 1989.

2.1-6: Hart, G.C. Jaw, J.W., Low, Y.K., SCM Model

for University .of JZolpxaAo. Flexural.

W

December 1989.

2.3-1: Ewing, R., Kariotis, J., El-Mustapha, A.,

LPM/I,._.A Computer Program tor _.the Nonlinear ,_ Dynamic Analysis of Lumped Parameter Models,

August, 1987.

2.3-2: Ewing, R., El-Mustapha, A., Kariotis, J.,

Influence of Foundation Model on the

Uplifting of Structures, July 1988.

3.1(a)-l: Scrivener, J., Summary of Findings of Cyclic

Tests on Masonry Piers, June 1986.

3.1(b)-l: Seible, F. and LaRovere, H., Summary of

Pseudo Dynamic Testing, February 1987.

3.1(c)-l: Herryman, K., Leiva, G., Antrobus, N.,

Klingner, R., In-Plane Seismic Resistance

of Two-Story Concrete Masonry Coupled Shear

Walls, September 1989.

1. TASK REPORTS ARE AVAILABLE THROUGH THE EARTHQUAKE

ENGINEERING RESEARCH LIBRARY, BERKELEY, CA.

95
3.2(a): Hamid, A., Abboud, B., Farah, M., Hatem, K.,

Harris, H., Response of Reinforced Block

Masonry Walls to Out-of-Plane Static Loads,

September 1989.

3.2(b)-l: Agbabian, M., Adham, S., Masri, S.,

Avanessian, V. Trama, Out-of-Plane Dynamic

Testing of Concrete Masonry Walls, Volumes 1 and 2, July 1989.

4.1- 1: Limin, H., Priestley, N., Seismic Behavior of

Flanged Masonry Shear Walls, Hay 1988.

4.2- 1: Hegemier, G. Murakami, H., On the_13ehayior of

Floor -to- Wall ^Intersections in Concrete

Masonry Construction: „ Part I: Experimental

4.2-2: Hegemier, G., Murakami, H., On the Behavior

of Floor-to-Wall Intersections in Concrete

Masonry Construction: Part II: Theoretical

6.2-1: Scrivener, J., Bond of Reinforcement in

Grouted Hollow-Unit Masonry: A State-of -theArt, June 1986.

6.2-2: Soric, Z. and Tulin, L. , Bond Splices in

Reinforced Masonry, August 1987.

8.1-1: Hart, G., A Limit State Design Method for

Rjeinforced Masonry.

9.1-1: Kariotis, J.C., Johnson, A.W., Design of

Reinforced Masonry Research Building,

September 1987.

9.2.2: Seible, F., Design and Construction of the

Charles Lee_ Powell Structural Systems

Laboratory, November 1986.

9.2.3: Seible, F., The Japanese Five-Story Full

Scale Reinforced Masonry Building Test,

January 1988.

11.1-1: TCCMAR, Summary Report: U.S. Coordinated

Program for Masonry Research, September 1985

to August 1986.

11.1-2: Status.Repor t : U.S. Coordinated Program for

Masonry Building Research, November 1988.

96
No. Topic

TABLE 3

TOPICAL REPORTS PLANNED

Pertinent Tasks

TRl

TR2

TR3

Material Properties & Tests 1.1, 1.2, 1.3

Reinforced Masonry Walls:

In-Plane Shear and Combined

In-Plane Shear and Vertical

Compression

Reinforced Masonry Walls:

Out-of-Plane Forces Combined

with Vertical Compression

TR4 Diaphragms

TR5 Bond and Splicing of

Reinforcement in Masonry

TR6 Limit State Design Concepts

for Reinforced Masonry

TR7 Modeling of Masonry

Components and Building

Systems

TR8 Large Scale Testing of

Masonry Building Systems

TR9 Determination of Earthquake

Induced Forces on Masonry

Buildings

2,1, 2.2, 3.1(a), 3.1(b)

3.1(c), 4.1, 9.4

2.4(a) , 2.4(b) , 3.2(a)

3.2(bl), 3.2(b2)

2.3, 2.4(a), 2.4(b),

Category 6 Tasks, 9.4

Category 3 Tasks, 4.1,

6.2

Categories 1-9

Category 2 Tasks, 3.1(b)

7.1, 9.4

3, Kb) , 3.1(c) , Category

9 Tasks

2.1, 2.3, 2.4(a), 2.4(b),

7.1, 8-1, 8.2, 9.4

97
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99
Design Guidelines of Medium Rise RM Buildings

by

Yutaka YAMAZAKI""'! , Shin OKAMOTO*^

Tsuneo OKADA*^, Akira MATSUMURA*'^ and Toshiyuki KUBOTA*^

SUMMARY

Design guidelines of medium rise RM buildings

were made by TECCMAR/Japan and were released March

1989. The design guidelines consist of eleven

chapters. They are: 1. General, 2. Quality of

Material, 3. Material Strength and Allowable

Stress, 4. Constants of Materials, 5. Structural

Planning, 6. Structural Calculation, 7. Allowable

Stress Design, 8. Total Flexural Resisting Moment,

9. Lateral Load Carrying Capacity, 10. Foundation,

and 11. Joint and Anchorage of Reinforcement,

Covering Concrete Thickness and Arrangement of

Reinforcement

.

In this paper, outline of these design

guidelines is presented.

KEY WORDS

RM Building, Design Guidelines of RM

Buildings, Seismic Design of Masonry Buildings

INTRODUCTION

The final goal of the U.S. -Japan coordinated

research on masonry buildings in Japan is to realize a medium rise reinforced masonry (RM)

building with less wall width ratio than the one required for present masonry structures. Design

guidelines for a medium rise RM building were made

by TECCMAR/Japan based upon the synthetic

experimental and analytical research which had

been carried out since 1984.

In the existing design codes for masonry

buildings in Japan, the number of stories is

limited up to three and a large amount of walls

(high wall width ratio) is required. In order to realize the earthquake resistant masonry buildings

which have larger story number but less wall width

ratio than the one required by the existing design

codes, a ductility design concept as well as a strength design concept was employed in the design

guidelines

.

The contents of the RM design guidelines are presented in APPENDIX II.

RM STRUCTURES AND THEIR REQUIRED SEISMIC

PERFORMANCE

RM Structures (Sections 1.1 and 1.4)

The RM structures which are dealt with in the

design guidelines are reinforced masonry

structures which are consist of reinforced masonry

(RM) bearing walls and wall girders, and

reinforced concrete (RC) slabs, foundations and

foundation beams.

The RM bearing walls and wall girders are to be fully grouted reinforced masonry systems which are made by stacking open-end type concrete or

clay RM units, as shown in Figs, la and lb, with

using joint mortar in parallel with arranging

reinforcements into unit cells vertically and

horizontally, as shown in Fig. 2.

r

\ ^

h^V^

f

18

1 75

'J

115

1?0 3^

Fig. la Concrete RM Unit

Scale Structure Testing Division,

Department, Building Research

Ministry of Construction, 1-

Tsukuba-shi, Ibaraki Prefecture,

Fig. lb Clay RM Unit

A RM building is basically to be an apartment

house type structure in which sufficient amount of

walls is expected to be provided in the transverse

direction

.

*1 Head, Large

Production

Institute

,

Tatehara,

#305 Japan

*2 Director, International Institute of Seismology

and Earthquake Engineering, BRI

""3 Professor, Institute of Industrial Science,

University of Tokyo, Roppongi, 7-22-1,

Minatoku, Tokyo, #106 Japan

*4 Professor, Dept. of Architecture, Kanagawa

University, Rokkakubashi , Kanagawa-ku,

Yokohama, #221 Japan

*5 Professor, Dept. of Architecture, Kinki

University, 3-4-1, Kowakae, Higashiosaka-shi

,

Osaka, #577 Japan

100
Fig. 2 Typical Bonding and Reinforcing Patterns

in the 5 story Full Scale RM Test Building

The RM structures shall be designed by either

"Design Method A" or "Design Method B" and should

satisfy the specifications listed in Table 1 for

each design method. In Table 1, the rigidity and

eccentricity factors are related to both vertical

and horizontal stiffness distribution limitations

in a structure and are defined in the Building

Standard Law [See Ref. 1]. The story drift angle

in Table 1 is defined as the one produced in each

story of a structure when subjected to design

lateral seismic shear forces which also are

specified in the Building Standard Law. The value

of the limitation, 1/2,000, for the story drift

angle was determined not to produce large story

drift angle, such as 1/200, against severe

earthquakes, say 300-400gals ground motion level

[Ref. 2]. Specifications on "shear stress of

wall" and "wall width ratio" are mentioned in the

following.

Table 1 Specifications for Design Methods A and B

for Design Method A for Design Method B

Shear Stress oE Wall

(kg/cm^)

T S 771/2

(3.1)

=f S 4 • JjJlSO

(5.5) Rigidity Factor (6.4) 2 0.3 2 0.6

Eccentricity Factor (6.4) S 0.3 S 0.15

Story Drift Angle (rad) (6.3) S 1/2,000 Standard Wall

Length Ratio

(for 1st Story)

(5.5)

5 stories 15cm/m2

3 stories 12cm/m2 Minimum Wall

Length Ratio

(for 1st Story)

(5.5)

5 stories 10cm/m2 12cm/m'^

3 stories 7cm/rii^ 9cm/m2

Story Height (1.1) £ 4.0m £ 3.5m

Building Height (1.1) S 16m S 16m

T ; shear stress of each wall (kg/cm ) t: nominal shear stress of walls in each story (kg/cm^)

F|^: specified compressive strength of RM assembrage

Rigidity and Eccentricity Factors: defined in Japan Building

Standard Law Story Drift Angle; story drift angle when subjected to design seismic force Wall Width Ratio: defined in each loading direction

Required Seismic Performance (Section 1.3)

1.3 Required Seismic Performance

(1) A RM building shall be a structure which

does not cause any trouble on serviceability against moderate

earthquakes and does not collapse in the event of severe earthquakes.

(2) A failure mechanism of a RM building

basically shall be governed by flexural

yielding at the base of walls at the

first story and wall girder ends.

(3) A RM building shall be a structure which

possesses the ultimate lateral strength

and def ormability corresponding to a structural performance coefficient

(symbolized as Dg hereafter) of 0.5 or more

.

Application

:

(1) Though a failure mechanism of a RM

building basically shall be governed by

flexural yielding at the base of walls

at the first story and wall girders,

flexural yielding at the top of walls

located at top story and a foundation

beam to which a wide wall is connected

may be permitted.

(2) Narrow walls may lead flexural yielding

or shear failure prior to flexural

yielding of wall girders to which the

nallow walls are connected. In this

case, it is needed to confirm that the

yielding or the failure of those walls

does not affect seriously on seismic

performance of the overall structure and

also that local collapse in the

structure does not occur. Similarly, if

there is possibility of shear failure in some wall girders, it is needed also to confirm that the failure of those wall

girders does not lead local collapse in

the structure.

The items (1) and (2) in main body in Section

1.3 will concretely signify that:

(i) shear cracking does not occur in most

structural components against design lateral

seismic shear forces corresponding to the

moderate earthquakes, and

(ii) a failure mechanism of a RM structure shall

be governed by flexural yielding which

certifies that almost no strength degradation occurs up to a story drift angle of

1/200 and that no major strength

deterioration does up to that of 1/100, as well as possessing sufficient lateral

strength against design lateral seismic

shear forces corresponding to the severe

earthquakes

.

DESIGN CRITERIOR •

' .

l.A Design Criterior

A RM building shall be designed both by

a working stress design method specified in

Chapters 6 and 7, and by an ultimate

101
lateral strength design method specified in

Chapter 9 as well as satisfying structural

specifications described in the following

Chapters except Chapter 8 (Design Method A). However, a RM building which satisfies the

special structural specifications described

in the following chapters shall be designed

without carrying out the ultimate lateral

strength design (Design Method B).

The RM structures can be designed either by

Design Method A or by Design Method B. In case of

Design Method A, seismic performance corresponding

to the value of 0.5 in the structural performance

coefficient D^ or larger is to be certified

through ultimate lateral strength design of an entire structure and ductility design of

structural components, whereas, in case of Design

Method B, seismic performance corresponding to the

value of 0.55_in the converted base shear

coefficient Cg or larger is to be certified

through calculating total flexural resisting

moment of an entire structure.

Limitation on the values representing

characteristics of a structure when applying

Design Method B is stricter than the one for

Design Method A as listed in Table 1, because in

Design Method B necessary amount of wall width and

reinforcement is determined based upon calculation

of flexural resisting moment without appropriately

assuming yielding mode of the structure, instead

of carrying out the ultimate lateral strength

design

.

The outline of Design Methods A and B is

represented in Table 2.

Table 2 Summary of Design Method

Design i^lechod A Design Method B

s a f J

:3 (T|_ S E3) S I,)

G -hsce n £ I .5

oOs • Ql

^h-re nSj.O

tfis ^ <3s.

(n Qr. tM^/1^)

oQs S aQ

oQs - Q,. * "I" where n^J.O

(fiS ^ '^su

(n Qj., r.Vlo'

o

Scory dri,fc An^l- s 1/2,000

Ulciroac- 3Crer>5:h and DtjctiiLC/ 0-3 Lgfl Ds 2 0.5Total ne<ura

D«3ign Cs

Re9L3Cing Moment

0.55

ea

^lu * Ol * f'-O- '-3' Omu

a

shear scresa du« to Lon^ cer^ d?3 coeEEicienc dec-r:nined according

Lgn load

to shear-span ratio, liaS2

1^CMOsu

allowable shear scress toe long cer-n design load

sh-iar fore- acctng on a me-nber dti* co lonj term design load shear force accmg on 3 .-ne^nber due co seiS'ULC design load

sum o£ yield Ele^ural tnomencs on boch ends o£ a .nember Eac^-co-Eacs span lengch of a (ne-nbfif coef£ici.enc co incrfase shear force shear force du"? Co shore Cerm design load allowable shear scrength

sum of fletural screngch on boch ends of a -ne-nbers

uIciiTiaC'- shear acrength

si'-ar force accing on a i-ne-nber ac mechan i s.ti coefficienc o£ ^crucCural pi^r for.-nanc

e

con verted base shear coefciclenc

5.5 Wall Width Ratio

(1) In each direction at each story, the

wall width ratio L,, in cm/m- which is

defined as the total width of bearing

walls in cm in each direction divided by

the floor area of the story considered

shall satisfy Equation 5.5.1.

L„ > a B

• Z and 5 Z • I^^

(5.5.1)

where

,

L.,„

2 standard wall width ratio in cm/m

indicated in Table 5.5.1 when the

thickness of walls t in cm is the

smallest thickness which is

defined in Section 5.3(2)

reduction factor calculated by

Equation 5.5.2 when the thickness of

walls t is greater than the smallest

thickness t which is defined in

Section 5.3(2)

I (tg X length of each wall)

1 =

Z (t X length of each wall)

a i ^ - (5.5.2)

reduction factor calculated by

Equation 5.5.3 related to the

specified compressive strength of RM

assemblage used

B =/l807T (5.5.3)

F^: specified compressi^ve strength of RM assemblage in kg/cm

Z : seismic hazard zoning coefficient

defined in the Building Standard Law^

L^j^: minimum wall width ratio in cm/m

indicated in Table 5.5.1

Table 5.5.1 Limitation of Wall Width Ratio

Scory from

Che Top (ground scory)

1-3

4 and 5

Scandard Wall Widch Racio Minimum Wall Widch Racio l^{cm/m )

Design Mechod A Design Method B

7

10

9

12

(2) The nominal shear stress when applying

Design Method B shall satisfy Equation

5.5.4.

T i 4.0 yF^/180

where

(5.5.4)

Cg = Q^/(Z

where

Q

z

R. W)

R.

ultimate lateral strength of a structure

seismic hazard zoning coefficient (= 0.7 - 1.0)

design spectral coefficient

shape factor = F

• Fg = 1.0 1.5) - W : structure weight

Fs (^e 1.0 1.5,

102
X : nominal shear stress in kg/cm^

produced in the walls in each

direction at each story when

subjected to the short term design

load corresponding to > 0.2

T = Q / Z

Q: seismic shear force in kg at each

story

Ay: sum of wall area in cm^ in each

direction at each story

F^: specified compressive strength of RM assemblage in kg/cm

(3) The calculation of the wall width ratio

shall be based on the following items

(i) through (iii).

(i) The bearing walls with the width more

than 59cm can be taken into account,

(ii) The bearing walls supported only by

wall girders can not be taken into

account

.

(iii) Full width of the bearing walls with

the small openings specified in

Section 5.4 can be taken into

account

.

The shear stress produced in a 5 story RM

building against the short term design load is

estimated as an example. ; Assume

:

wall width ratio: 15cm/m

shear stress due to the long term design load: 0

wall thickness: 19cm

story weight: l,200kg/m^

seismic hazard zoning coefficient: 1.0

Then, nominal shear stress is calculated as

0.2 X 1200 X 5 T =

15 X 19 = 4.21 kg/cm2

Taking shear stress concentration coefficient

n into account, shear stress produced in a critical (wide) wall in the story becomes;

T = n • T = 6.32 kg/cm2 (p = i .5)

6.74 kg/cm2 (n = 1.6)

The allowable shear stress fg for the short

term design load in case of F^ = ISOkg/cm^ is

obtained based on the equation, = JY~/2

specified in Section 3.1 as;

fg = 6.71 kg/cm^

This example shows that shear stress produced

in most walls is smaller than the allowable shear

stress level and that no shear cracking is

expected to be produced in most walls in RM

buildings having standard weight against the short term design load corresponding to Cq = 0.2.

QUALITY OF MATERIALS

RM Units (Section 2.1)

Masonry units for RM buildings shall conform

to the "Standard of Quality of Masonry Units for

Medium Rise RM Buildings". In the Standard, units

are classified into two kinds according to their

materials made from, namely, concrete and burned

clay (or ceramics). Concerning dimensions of

units, these are also assorted into two types,

such as "12" type and "23" type. The "12" type has

its sizes with the ratio of thickness to length as

1 : 2. The "23" type has the ratio of 2 : 3. Combinations of dimensions have variation in some

range, as listed in Table 3.

Table 3 Dimensions" of RM Units

Type Thickness(mm) Length(nim) Height(mm)

200 400

12 225 450 100, 150, 200

250 500 225, 250, 300

300 600

200 300 100, 150

23 266.7 400 200, 225, 300

300 450

Dimensions are expressed in nominal sizes, i.e.,

distances from center to center of the thickness

of morter joints.

The Standard also classifies the quality of

units into three grades according to their

compressive strength and water absorption. Table

4 denotes its classification. Difference in

strength values between units of concrete and clay

in the same grade comes mainly from their

conformity of water absorption values in each

grade.

Table 4 Strength and Water Absorption

Material Grade

Compressive

Strength, min.

kg/cm

Net Area

Water

Ratio

Absorption

max. % vol.

1 400 10 Concrete 2 300 15

3 200 20

Clay

123

600

500

400

10

15

20

Specified Compressive Strength (Sections 2.2, 2.3 and 2.5)

Specified compressive strength of grout

concrete and mortar shall be more than 180kg/cm . Compressive strength of joint mortar shall be high

enough to obtain the specified strength of RM

assemblage. Specified compressive strength of

concrete which is used in R/C portion shall be

ranged between 180kg/cm and 270kg/cm . Specified compressive strength o;^ RM

assemblage shall be ranged between ISOkg/cm"" and

270kg/cm".

103
Materials of RM Assemblage (Section 2.4)

Specified conpressive strength of RM

assemblage shall be verified by prism testing as a rule

.

Instead of applying prism testing, specified

compressive strength of RM assemblage may be

verified by using Equation A 2.4.1.

where

= eg

I (l-B')F^ + B'Fg } (A 2.4.1)

= specified compressive strength of RM

Assemblage (kg/cm^)

= specified unit strength (kg/cm^)

= specified grout strength (kg/cm'^)

y = cavity ratio of unit

= masonry reduction factor, which value is 0.75 as a rule. However when strength of

joint mortar is less than F^j, it shall be ratio of the former to the latter, and

not larger than 0.75.

Specified compressive strength of grout shall

be more than that of RM assemblage as a rule.

Allowable Stresses of RM Assemblage (Section 3.1)

o

3.0

2.0

1.0

E =1.31xl05 yFm/180

• Concrete RM

O Clay RM

o^^'^oo' 300 400

2> Prism Strength F^.^^ (kg/cm^^)

Fig. 3a Young's Modulus of Elasticity

vs. Prism Strength

Allowable stress of RM assemblage shall be in

accordance with the values in Table 5.

Table 5 Allowable Stress of RM Assemblage

Description

Allowable Stresses (kg/cm^)

For Stresses due

to Long Term Design Load

For Stresses due

to Short Term

Design Load

Compression Fm/6 Fm/3

Shear

Modulus of Elasticity of RM Assemblage (Section

A.l)

Young's modulus of elasticity shall be

obtained by compression prism test as a rule.

It may also be estimated by Equations A4.1.2a and

A4.1.2b, rather conservatively (Fig. 3a).

Young's modulus of elasticity:

for concrete masonry^

Em = 1-68 X 10^^^

for clay masonryE„ = 1.31 X 10^ 7:

/L80 (kg/cm")

2^ F^/180 (kg/cm"^)

(A4.1.2a)

(A4.1.2b)

Eq. A4.1.3

•Concrete RM >Clay RM

U 1 2 Eg6'/E^(l-6')

Fig. 3b Eq. A4.1.3 and Test Results

If moduli of elasticity of constituent materials are known, Young's modulus of elasticity of

masonry may be estimated by using Equation A4.1.3 (Fig. 3b).

= d-e') Eu +1 (kg/cm2) (A. 4. 1.3)

Where E^ = Young's modulus of elasticity of masonry

E^ = Young's modulus o£ RM unit

Eg = Young's modulus of grout

ot= cavity ratio of unit

Shear modulus of elasticity is able to be

estimated by Equation A4.1.1 using a value of

Poisson's ratio measured in compression prism test.

(A4.1.1)

mwhere

2( 1+ V m)

^m = shear modulus of elasticity of masonry

= Poisson's ratio of masonry:

104
STRUCTURAL SPECIFICATIONS FOR RH

COMPONENTS

Bearing Walls

5.3 Bearing Walls

(1) The bearing walls shall be greater

than 59cm in width.

(2) The minimum thickness of bearing walls

shall be greater than 19cm and 1/20 of

the vertical distance between main

lateral supports.

(3) The bearing walls of the upper story

shall be located above those of the

adjacent lower story as a rule.

(4) The bearing walls shall have no openings as a rule. However, the small

openings regulated in Chapter 5.4 may

be provided.

(5) The vertical and horizontal shear

reinforcements conforming to Section

7.2, Chapters 9 and 11 (Section 7.2,

Chapters 8 and 11 for Design Method B)

shall be provided in the bearing walls.

7.2.4 Arrangement of Reinforcements in the

Bearing Walls

(1) Vertical and shear reinforcements shall

be arranged with those more than the

required amount of reinforcements

determined from the allowable stress

design and specified amount of

reinforcement sin Table 7.2.1.

Table 7.2.1 Reinforcements in Bearing Walls

Scory from

Che Tap

End Flexural

Reinforcetnenc

Incecmed Lace VercLcal Reinforcement

Verctcal Reinforcemenc

Racio

Shear Reinforcemenc

from 1 Co

2

2:l-Dl6 and i2-D25

Space S Lengch of Unic 2:0.22

Racio 20.21 (0.251) Space SHeighc of Unit

S20cta

from 3 Co

5

Racio 2G.25Z C0.3Z) Space ^Heighc of Unic

or

Size of Reinforceaenc

D16-D25 Dl0-Dt6 D10-D16

Horn: Values in parencheses denoce chose for Design Mechod B.

(2) Surroundings of the small openings

specified in Section 5.4 shall be

properly reinforced by arranging 1-DlO

reinforcement or larger ones.

(3) Number of reinforcements which could be arranged in a unit hole shall be less

than 2. However, the number of

reinforcements shall be less than 4,

if the thickness of RM units is thick

enough to arrange the reinforcements in

two lines.

Wall Girders

5.6 Wall Girders

(1) Wall girders shall be provided to effectively connect bearing walls each

other at every story.

(2) The end of wall girders shall be

supported by a bearing wall (including

an intersecting bearing wall).

(3) Wall girders shall have sufficient

capacity against both vertical and

horizontal design load, and shall

satisfy the following items (i) through

(v).

(i) Width of wall girders shall be more than the thickness of a bearing

wall connected to them,

(ii) Depth of wall girders shall be more

than 45cm.

(iii) The ratio of clear span 1^ in cm to depth D in cm of wall gir de r s ,

1

q /D,

shall be more than 1.5. (iv) The lateral and shear reinforcements

conforming to Section 7.3, Chapters 9 and 11 (Section 7.3, Chapters 8 and 11 for Design Method B) shall be provided

in the wall girders.

7.3.4 Arrangement of Reinforcements in Wall

Girders

(1) Lateral and shear reinforcements shall

be arranged with those more than the

required amount of reinforcements

determined from the allowable stress

design and specified amount of

reinforcements in Table 7.3.1.

Table 7.3.1 Reinforcements in Wall Girders

Ejid Fletural Reinforcemenc Incermediace

Laceral

Rein forcemenc

Laceral Reinforcemenc

Racio

Shear Reinforcemenc

AmounC of Reiriforcenencs i 1-D16

and S2-D25

/ 1-D19 \

( and 1 ^ ^2-025''

SO. 251 ao.25Z

(2:0. 3Z)

Space 2; 20cin

Size of ReinforcemencD16-D25 D10-D25 DlO-Dt6

Noce: Values in parencheses denoce chose for Design Mechod B-

(2) In case of short beams, the end flexural

reinforcements shall be more than 1-D16

reinforcement and the shear

reinforcement ratio shall be more than

0.3%.

(3) Number of reinforcements which could be arranged in a unit hole shall be less

than 2.

Floor Slabs

5.8 Floor Slabs

(1) Floor slabs as a main structural

component shall be constructed by

reinforced concrete and have sufficient

capacity against vertical load. They

shall also have sufficient strength and

rigidity to transfer the stress caused

by lateral force to bearing walls and

wall girders (or to foundation beams at

the lowest story).

(2) Thickness of the floor slabs shall be more than 13cm.

105
Foundations and Foundation Beams

5.10 Foundations and Foundation Beams

(1) Foundation beams shall be effectively

and continuously provided beneath the

bearing walls at lowest story.

(2) Foundations and foundation beams shall

have sufficient strength and rigidity

against vertical and lateral load, and

satisfy the following items (i) though

(v).

(i) Foundations and foundation beams shall

be cast-in-place reinforced concrete,

(ii) Width of foundation beams shall be more than the thickness of bearing

walls connected to. (iii) Foundation beams shall be reinforced

at four corners in their section

against flexural moment. The flexural

reinforcements shall be restrained by

shear reinforcements,

(iv) Foundations shall be properly arranged

beneath the bearing walls as a rule,

(v) Foundations and foundation beams shall

have sufficient resistance against

torsional stress produced due to eccentric arrangement of foundations

and/or piles.

7.7 Foundation Beams

(1) Design stress for foundation beams shall

be calculated based on Section 7.1. (2) Design of foundation beams shall be

carried out in accordance with AIJ R/C

Standards

.

(3) Foundation beams shall satisfy the

following items (i) through (iii).

(i) The amount of both upper and lower

flexural reinforcements shall be more than 2-D16 as a rule.

(ii)The spacing of shear and

intermediate lateral reinforcements

shall be less than a half of the depth

of foundation beams and shall be less

than 30cm.

(iii) Shear reinforcement ratio shall be more than 0.2%.

Bearing Wall-Wall Girder Joints

5.11 Bearing Wall-Wall Girder Joints

(1) The thickness of bearing wall-wall

girder joints shall be more than the one of adjacent bearing walls and wall

girders

.

(2) Small openings shall be permitted to be provided for the joints to which a bearing wall having the width 1 more

than 159cm is connected.

(3) Small openings in the joints shall

satisfy the following items (i) through

(iv).

Bearing Wall-UalL Girder Joinc

^5^^^;^ Opening

Wall Girder

Floor Slab

f i ) ^ 2 1 5 9 cm

ii ) R , . R I S 2 0 cm

iH) I , S 3 Oct. and f , S D,/2

iv) it S 3 Oca and 2 I 2 D z/

2

v) I, S (R, + R,)

vi) D, SD,/2 and Da S D 2

Fig. 5.11.1 Small Openings in Bearing

Wall-Wall Girder Joint

(i) Clear length R (to be the diameter for

round shape and the longer side for

rectangular shape) shall be less than

20cm.

(ii) The distance l^^ and I2 between the

side of bearing walls and the edge of

small openings shall be more than

30cm and a half of the depth D^ and D2

of wall girders connected to. (iii) The plural small openings in the joint

shall be arranged side by side, and

the distance between the edges of two

small openings 1-^ shall not be less

than the sum of clear length R-|^ and R9 of those small openings as a rule.

(iv) The small openings shall not be

located at a lower half portion of the

joints

.

7.8 Bearing Wall-Wall Girder Joints

(1) The long and short term allowable stress

design of the bearing wall-wall girder

joints shall be abbreviated to be

carried out. However the examination

specified in Section 9.5.4 shall be

needed for the joints which satisfy the

following items in the whole, whether

Design Methods A or B is applied.

(i) The width of bearing walls to which

the joint is connected is smaller

than 98cm.

(ii) The joints to which the wall girders

are connected in the both side

(iii) The amount of tension side flexural

reinforcements in the wall girders

stated in the above item (ii) is more

9 than Ocm'-.

(2) The surroundings of small openings in

the joints which are specified in

Section 5.11 shall be reinforced in

accordance with the specifications in

Section 11.4.3.

Seismic Slit Joints and Shear Keys

5.12 Others

(1) Seismic slit joints shall be provided

between bearing walls/wall girders and

secondary walls in order not to have

bad influence on strength and

deformation capacity of those bearing

walls/wall girders.

106
(2) Shear keys shall be provided as a rule,

in order to fully transfer the shear

forces, on upper surface of foundation

beams and floor slabs on which the

bearing walls are arranged.

STRESS ANALYSIS

6.2 Stress Analysis

(1) Stress and deformation of RM buildings

shall be evaluated using elastic

stiffness of structural components as a rule, in which;

(i) Stress under the vertical design load can be obtained applying abbreviated

methods which are practically applied

under actual circumstances,

(ii) The long term axial stress acting on the bearing walls can be evaluated

against the vertical design load

subjected to the floor area surrounded

by bisectional lines which devide

corner angles at bearing wall

intersections, clear distance between

two parallel bearing walls, and clear

length of wall girders.

(2) The effective width of orthogonal walls

and floor slabs shall be appropriately

taken into account to evaluate stiffness

of T-shape structural components.

(3) The effect of torsion shall be taken

into account if the center of gravity

and the center of rigidity at each story

do not coincide with each other.

Application

:

(1) Stress and deformation of RH buildings

can be evaluated using a frame analysis

method in which flexural and shear

deformation and rigid zone at bearing

wall-wall girder joints are taken into

account. Axial deformation also shall

be taken into account for those bearing

walls in which large axial deformation

is expected. The rigid zone area taken

into account is determined based on the AIJ R/C Standards.

(2) The stiffness of foundation beams shall

be taken into account. Deformation of

the ground and piles can be

appropriately evaluated as well.

(3) Young's modulus and elastic shear

modulus of RM structural components

shall be those for the RM assemblage

specified in Section 4.1. The effect of

existence of reinforcements may be taken

into account as well.

(4) The plural bearing walls which surround

openings having the value of an equivalent opening ratio of 0.4 or less

can be considered as an equivalent non- opening single bearing wall. The shear

stiffnesses in this case shall be

evaluated in accordance with the shear

stiffness reduction factor

specified in the AIJ R/C standards

(Eq. A 6.2.1).

Shear stiffness reduct ion factor Y

Y = 1 - 1.25 / (A 6.2.1)

hi

where

o o 0.4

hi

and

h: vertical distance between main

lateral supports of bearing walls

1: width of bearing walls

h^, 1^: clear height and width of the

opening, respectively

(5) The reduction of shear stiffness may be

neglected for the small openings

specified in Section 5.4. (6) The flexural stiffness of wall girders

in which clear length is comparatively

short (designated as "short beams") may

be reduced up to 50% of their elastic

stiffnesses

.

TOTAL FLEXURAL RESISTING MOMENT

IN DESIGN METHOD B

.1 Check of t+ie Total 'Flexural Resisting

Moment

The following Equation 8.1.1 shall

be satisfied in the case that design

method B is used.

(ZMbu + Z,Mwu) ^ 0.55 • Z • Rt

• Z (Ai • Wi • hi) (8.1.1)

i=1 where

,

EMbu: total summation of moment

capacity of wall girders at nodal points (kg cm)

ZiMwu: total summation of moment

capacity of walls at the bottom

of 1st story (kg cm)

Z : seismic hazard zoning

coefficient

Rt : design spectral coefficient

Ai : lateral shear distribution

factor at i-th story

Wi : weight of the building above the

i-th story (kg)

hi : height of the i-th story (cm)

n : number of story

For the buildings to which design method B is

adopted, the structural regulations such as wall

width ratio, profile of plan and elevation, etc. are required strictly. However, the state at ultimate stage of the structure is not clear

because only allowable stress design method is

used . In this section the simplified check of the

lateral load carrying capacity is obligated. The

mechanism as shown in Fig. 4 is assum.ed whithout

checkinig the yielding mechanism in the

calculation

.

107
Fig. U Yield ing Mechanism assumed

in the Design Method B

LATERAL LOAD CARRYING CAPACITY

Fundamental Concept (section 9.1)

9.1 Fundamental Concept

(1) The lateral load carrying capacity of

RM buildings shall be greater than the

required lateral load carrying capacity

expressed by Eq. 9.1.1.

Qun = Ds * Fes • Qud

where

,

Qun

(9.1.1)

(2)

(3)

(M

required lateral load carrying

capacity at each story

Ds : structural performance

coefficient specified in Section

9.2 Fes : shape factor which is determined

based on the eccentricity and

variation of lateral stiffness

along the building height

Qud : lateral seismic shear force at each story

Failure mode of members shall be

flexural yielding mode as a rule.

Failure mode of a building shall be

total failure mode in which flexural

yielding occurs at the bottom of the 1st story walls and at the ends of wall

girders as a rule.

Walls, wall girders, foundation beams

and surroundings of small openings shall

be reinforced according to the

requirements specified in Section 9.5.

The procedure of the check is as follows;

(i) to calculate ultimate strength of each

component

,

(ii) to calculate lateral load carrying

capacity in every direction and to decide

the stress and yield type for each

component under yielding mechanism of a building

,

(iii) to decide the structural performance

coefficient "Ds" at the direction

considered using the "component type" for

wall girders and bearing walls stipulated

in Section 9.2

,

(iv) to calculate the required lateral load

carrying capacity at each story and

compare with the lateral load carrying

capacity obtained from item (ii).

(v) to do the ductility design for bearing

walls, wall girders, foundation beams and

surroundings of small openings according

to the requirements specified in Section

9.5.

Structural Performance Coefficient (Section 9.2)

9.2 Structural Performance Coefficient Dg

(1) The structural performance coefficient

,Dg, shall be the maximum value which is

defined in Table 9.2.1 for walls

according to the component types.

Table 9.2.1 Structural Performance Coefficient

Ds

in Che case thac only F]^ component

IS considered for

the calculation

of lateral load carrying capacity

0.5

in the case that

F]^ and F2 components are considered for the calculation of lateral load carrying capacity

0.55

in the case that

Fj , F2 and F3 compomponents are considered for the calculation of lateral load carrying capacity

0.6

(2) The judgment whether Fl, F2 or F3 is

selected may be performed according to the conditions in Table 9.2.2 a) and b)

in which the conditions are stipulated

for walls, wall girders and foundation

beams

.

Table 9.2.2 a) Component Types of Bearing Walls —...Component Type Conditions —

Fl F2 F3

Failure Mode Flexural 1 "allure*^" other Upper Limit of lu/Fm 0.07 0.10 conditions

Upper Limit of Ooe/Fm 0.2 0.25

Table 9.2.2 b) Component Types of Wall Girders

and Foundation Beams

Com ponen ts

-~.-Com ponen t Type Conditions ......^

Fl F2 F3

Wall Girders

Failure Mode Flexural Failure*

other

Upper Limit conditions

of Tu/Fm

0.07

Foundation BeamsFailure Mode Flexural Failure*

other Upper Limit conditions

of Tu/Fm 0.15 0.20

* The flexural failure implies that the ratio of the shear capacity to the shear at mechanism is greater than 1.1 for walls and

1.0 for wall girders and foundation beams, respectively

.

Furthermore, the ratio of the shear

capacity to the shear at mechanism

should satisfy the requirements of Section

9.5.

Tu: average shear stress

mechanism (kg/cm'^)

Tu = Qmu / Ae (9.2.1)

where,

Qmu : shear at mechanism (kg)

a t

108
Ae : sectional area of wall or wall

girder in which the sectional area

of slab and transverse wall within

the effective range specified in

Section 9.4 may be included, (cm^)

Ooe : average axial stress of

bearing wall at mechanism (kg/cm^)

aoe = ( Nm + Nw ) / Ae (9.2.2)

where

,

Nm : Nw

Ae

axial force at mechanism (kg)

axial force from transverse wall

which affects the strength of the

bearing wall and is same as the

value of Section 9.4.1 sectional area of bearing wall in

which the sectional area of

transverse wall within the

effective range may be included

Fm : specified compressive strength of RM assemblage

Tu : shear stress of foundation beam at mechanism (kg/cm )

Tu = Qmu / ( b • j ) (9.2.3)

where

,

Qmu

b

j

shear of foundation beam at mechanism (kg)

width of the foundation (cm)

distance between the centroids of

compressive and tensile force at the section (cm)

(3) In the case that yield hinges occur at wall girders which are connected to

a bearing wall at top and bottom, the

lowest level of component type among the

component types of wall girders shall be

selected for the bearing wall.

(1) Structural Performance Coefficient

From the response analysis of the building

designed by these guidelines, the response of

story drift angle was less than 1/200 under the 50

kine input of Hachinohe 1968NS, Taft 1952EW and El Centro 1940NS. Furthermore, the component test

indicates that the bearing wall with medium amount

of shear reinforcement and low level axial index ( Oo/Fcm) has def ormability of more than 1/130 story

drift angle (Fig. 5). In case of wall girder it was more than 1/100 (Fig. 6). Therefore, the

coefficient of structural performance is decided

to be 0.5 for the structure with ductile members,

0.6 for the structure including brittle members

and 0.55 for the middle. The reason that the

simple summation is allowed for the structure with

various component types is such that the yielding

deformation at maximum load was almost 1/400 for most of the tested members.

(2) Component Type

The component type may be classified by the

factors: the deformability such as failure mode,

axial force index ( aoe/Fm ), shear-span-ratio ( M/QD ), level of shear stress, amount of shear

reinforcement, etc., which affect the

deformability. For the classification of component

type, the factor of failire mode is stressed

first, because the earthquake response of this

kind of structure is rather small and the

structure has the deformability of 1/200 to 1/100

in story drift angle if the failure mode is

flexural. However, it seems to be difficult to give enough deformability to the structure under

high axial and/or shear stress. Hence the

limitation for them was provided.

For the bearing walls the relation between

the limit of deformation angle, which is defined

when the load is decreased 20% after maximum load,

and shear stress index ( Tu/Fm ) is shown i^ Fig.

7. The deformation angle more than 7.5x10 rad.

is expected when the T^/Fm is less than 0.07. For

the wall girdirs the relation is shown in Fig. 8. The deformation angle more than 10x10 rad. is

expected when Tu/Fm is also less than 0.07.

o 0.025

0.02

0.015

0.01

0.005

0

G O

O

C

u

O 0 C 0

•

-^^^^

c ^1

••

3

0.5 1.0 1.5 2.0

Qsu / Qmu

Fig. 5 Deformabiliy of Bearing Wall

Symbols: O flexural failure, concrete unit ©shear failure after flexural yield, concrete unit # shear failure, concrete unit

flexural failure, clay unit 0 shear failure after flexural

failure, clay unit H shear failure, clay unit A flexural failure, with transverse

component A shear failure after flexural failure, with transverse component X shear failure, with transverse

component

0.03

0.02

0.01

®

oeg 1 ^

§ o

• a ->- Short Bearn

0.5

Qsu / Qmu

Fig. 6 Deformability of Wall Girder

109
O

00

4-J to o

OJ CO

r-l o

00 rH C

CO 3 c E o H

•H X

4J CO

CO E E

i-1 o

4-1 o

<u Q

20

15

10

o

k

2

•

• ^ > G

•

Fii

0 0.C5 0.07 O.fO 0.15 0.20 Tu / Fcm

. 7 Shear Stress and Def ormability

of Bearing Wall

6-9 O

CO

at -a

cu COo

r—

1

c

CO E

c E o H H X

4-) CO CO E e

U CM o o

aj Q

0.05

0.04

0.01

(Ll

_L

Fig,

0.05 0.10 0.15 Tu / Fcm

8 Shear Stress and Deformability

of Wall Girder

Calculation Method for Lateral Load Carrying

Capacity (Section 9.3)

9.3 Calculation Method for Lateral Load

Carrying Capacity

(1) Accurate calculation method such as incremental load applying analysis or simplified method based on virtual work

principal shall be used for the

calculation of lateral load carrying

capacity

.

Though the incremental load applying analysis

or the limit analysis is desired for the

calculation of the lateral load carrying capacity,

the simplified method based on virtual work

principal assuming the yielding at the ends of

wall girders may be used because adequate accuracy

may be obtained for the structures in which the

strengths of walls are bigger than those of wall

girders. I'he moment distribution method is not suitable for this kind of structures. The moment and shear force for each component

calculated by the above mentioned method may be

used as the moment and shear at mechanism.

However, for wide walls the forces calculated by

the simplified method are sometimes different from

those by the accurate method. The increase of

design shear forces for the wide walls is

reccommended if the simplified method is used for

the calculation.

Ultimate Strength of Components (Section 9.4)

9.4.1 Bearing Wall

(1) The Eq. 9.4.1 shall be used for

fle.xural strength of bearing walls.

the

M u = Z ( a t • + 0.5 • Nmay ) Iw + 0.5 ZCaw

• lw + Nw» e (kg

0 w y ) 1 w cm) (9.4.1)

where, Mu = 0 when it is negative,

at: sectional area of flexural

reinforcement in which the

vertical reinforcement sectional

area of transverse wall within the

effective range may be included

(cm2)

oy : material strength of flexural

reinforcement specified in Section

3.3 (kg/cm2)

Iw : 0.9 times the total length of the

bearing wall (cm) aw : sectional area of vertical

reinforcements at the center part

of the bearing wall in which the

reinforcement sectional area of

transverse wall within the

effective range may be included

(cm2)

owy : material strength of the above

mentioned reinforcement specified

in Section 3.3 (kg/cm ) Nm : axial force of the bearing wall at mechanism (kg)

Nw : axial force from the transverse

wall which affects the strength of

bearing wall

In the case of the wall as shown in

Fig. 9.4.1 the calculation shall be as follows.

i) Nw =0.25 N| for the bearing wall

Wl ii) Nw =0.25(N^ + N2) for the

bearing wall W2

iii) Nw =0.25 Nn for the bearing wall

W3

Nl and N2 are the long term load

(kg) of the transverse walls.

Transverse wall

y ^ Bearing wall Wl r

Bearing wall W2

•Bearing wall W3

Fig. 9.4.1 Axial Force of Transverse Wall

110
the distance (cm) between the

compression extreme fiber of

the bearing wall and the center

line of Che transverse wall

(Fig. 9.4.2)

Tension Compression

side e side

9.

Compression fiber

Fig. 9.4.2 Distance Between Compression

Extreme Fiber and Centroid of

Transverse Wall

(2) The Eq. 9.4.2 shall be used for the

shear strength of bearing walls.

Qsu =

where

,

te

^ 0.053pteO-23 (Pm +180)

1 M /( Q • 1 )+0.12

+0.1 ooe [te • j • r

+2.7 y pwe-Owy

(kg) (9.4.2)

p te

at

Fm

Iw

pwe :

Ow y

: equivalent thickness when total

sectional area Aw including the

the sectional area of transverse

wall within effective range is

replaced by the rectangular

section with the total length 1 (cm)

This shall be less than 1.5 times the thinkness of the

bearing wall.

equivalent flexural reinforcement

ratio ( = 100Zat / (te • Iw )) (%)

sectional area of flexural

reinforcement in which the

sectional area of transverse

walls within the effective range

may be included (cm^)

specified compressive strength of

RM assemblage (kg/cm^)

total length of the bearing wall

(cm)

0.9 times the total length of the

bearing wall (cm)

equivalent shear reinforcement

ratio (= a^/(te • s))

The "a^" is sectional area of one

pair of shear reinforcement and the

"s" is the space of it. When this

value is more than 0.012t/te it

shall be 0.012t/te.

material strength of shear

reinforcement of bearing wall

( ^3000 kg/cm2)

(8/7) Iw (cm)

reduction factor of shear capacity

due to small openings, which value

shall be the smaller value of

l-lo/l and 1- ho lo

h 1

ho

lo

a oe

vertical distance between the main

lateral supports in the bearing

wall (cm)

height of small opening (cm)

width of small opening (cm)

ZN/ ZAw

ZN= Nm +Nw

Nm : same value to that in Eq. 9.4.1 Nw : same value to that in Eq. 9.4.1 Aw : total sectional area of bearing

walls including transverse walls

within the effective range (cm ) H/(Q'2,): shear span ratio at mechanism

which shall be 1 and 3 when it

is less than 1 and more than 3, respectively

M and Q are the bending moment

and shear force of the section

at mechanism.

(1) Formulae for flexural strength

Figure 9 shows the relationship between the

experimentally obtained shear stress at maximum

load, J.T max =Qmax/t • 1, and shear stress

calculated by Eq. 9.4.1, (-Xmu = ZcMu/t-1, whose

data were obtained from 12 specimens, which were failed in flexure, among 57 specimens in total.

The test data are almost same as the calculated

value by the equation. In this case the results of

material test were used for the yield strength of

reinforcement. In these guidelines the equation by

which the average strength can be calculated is

adopted, because the design of each component

against shear failure becomes critical if the

equation which gives lowest value is used.

(2) Effect of transverse wall on flexural strength

The test results of bearing walls with

transeverse walls show that all vertical

reinforcement in transverse walls yielded when the

deformation angle was 1/200 to 1/100. The fullscale five story test showed that total sectional area was effective at final stage. Since the

target of the deformation angle in these

guidelines, however, is a little smaller, 1/400 to 1/300, than 1/200 to 1/100, the effective range is

limitted in the requirement.

(3) Shear strength

Figure 10 shows the relationship between the

experimentally obtained shear stress at maximum

1 oa d , J. T max = Qm a X / t • 1, and shear stress

calculated by Eq. 9.4.2, ^Tsu=cQu/t«l , whose data

were obtained from 45 specimens, which were failed

in shear or failed in shear after flexural yield,

among 57 specimens in total. All data are plotted

above the diagonal line. The equation seems to give the lowest limit. The material test results

are used for the calculation above.

(4) Effect of transverse wall on shear strength

There are 3 specimens with transverse walls.

The wall with transverse walls at the center

failed in shear after flexural yield and the wall

111
with transverse walls at the ends failed in shear.

The ratio of the strength of the wall with

transverse walls to that without transverse walls

is 1.43 in average and the ratio of the strength

with transverse walls to that without transverse

walls calculated by Eq. 9.4.2, in which te is

assumed to be 1.5, is 1.36 in average.

0 15 30 shear stress at flexural strength

(calc. fmu, kg/cm^)

9 Shear Stress at Maximum Load versus Shear Stress at Flexural Strength

40

20

'a

I K

1 / i /

/ 1

1

1 «

' H » 1

T /s- A

1

1

1

1

1

1

1 Shear Failure

1

1

1 0 20 40

Fig.

shear stress at flexural strength

(calc . T mu , kg/cm ) 10 Shear Stress at Maximum Load versus

Shear Stress at Shear Strength

9.4.2 Wall Girders and Foundation Beams

(1) The Eq. 9.4.1 shall be used for the

flexural capacity of wall girders.

Mu =0.9 • i:(at • oy . d) (kg • cm) (9.4.3)

where

,

at

ay

sectional area of flexural

reinforcement- in which the

sectional area of the slabs within

the effective range shall be

included (cm-)

material strength of the flexural

reinforcement specified in Section

3.3 (kg/cm2)

d : effective depth of the wall girder

(refer to Section 7.3.2) (cm)

(2) The Eq. 9.4.4 shall be used for the

shear strength of the wall girders.

Qsu =0.9'

f0.053 pte°-23 (Fm +180 )

M / (Q

2.7 /p w e

• d) + 0.12

Ow y

I be J (kg)

(9.4.4)

where

,

be equivalent width of the wall

girder when total sectional area of the wall girder

including the sectinal area of the slab within the effective

range ZAg is replaced to the

rectangular section with depth D (cm)

This value shall be less than 1.5 times the width b of the wall gider.

be = Z Ag / D, be S 1.5 b

equivalent flexural reinforcement

ratio

(= lOOZ at / ( be . d ) ) (%)

sectional area of flexural

reinforcement in which the

sectional area of slab reinforcement

within the effective range may be included (cm^)

specified compressive strength of RM assemblage (kg/cm'^)

depth of the wall girder (cm)

equivalent shear reinforcement ratio

(=a^/(be.s))

The a,," is the sectional area of a

w pair of shear reinforcement and

the "s" is the space of it. When

this value is more than 0.012 b/be,

it shall be 0.012 b/be.

Owy : yield point of shear reinforcement

of the wall girder (< 3,000 kg/cm^)

M/(Q'd):shear span ratio at mechanism which

shall be 1 and 3 when the value is

less than 1 and more than 3,

respectively

M and Q are bending moment and

shear force of the section at

mechanism

.

j : (7/8)d may be used.

(3) The flexural strength of the foundation

beam may be calculated by Eq. 9.4.3. The shear strength of the foundation beam

may be calculated by Eq. 9.4.4 in which

the reduction factor 0.9 in the equation

may be changed to 1.0.

pte

a t

Fm

D pwe

(1) Flexural strength and shear strength of wall

girder

Equations 9.4.3 and 9.4.4 were examined using

the test results of 33 specimens in total.

The comparison of the test results and the

calculated values by Eq. 9.4.3 is shown in Fig.

11. The specimens, plotted under the diagonal

112
line which shows the ratio of test results to calculated value is 1.0, were failed in shear.

For the specimens failed in flexurure, the ratio

is 1.2 to 1.5. The comparison of the test results and the

calculated values by Eq. 9.4.4 is shown in Fig.

12. The specimens, plotted under the diagonal line were failed in flexurure. For the specimens

failed in shear, the ratio is 1.1 to 1.5. Equation

9.4.4 seems to be coservative. The failure modes

are clearly divided by the diagonal line as shown

in Fig. 13.

30

25

20

15

0

1 /

•

(D

/

10 -

T

0 5 10 15 20 25 30

shear stress at flexural strength

n (calc Tmu, kg/cm ) Fig. 11 Maximum Strength versus Flexural

Strength of Wall Girder

00 30

25

20

15

10

'•/ a\

0 10 15 20 25 30

shear stress at shear strength

(calc fsu, kg/cm2)

Fig. 12 Maximum Strength versus Shear

Strength of Wall Girder

u

XI

4-1 00

Gfl C

OJ

4-J

C/l \t\ H

1—

1

C33 a

X

<X> r-\

at

cnm

<uu

4-1 OTu

CO

(U ^

cn

30

25

20

15

10

[

1 1

- ': ! -J^- ^/ =

/ i i i 10 15 20 25 30

shear stress at shear strength (calc Tsu, kg/cm^)

Fig. 13 Flexural Strength versus Shear

Strength of Wall Girder

Ductility Design of Components (Section 9.5)

9.5.1 Walls

(1) For the component type Fl and F2 of

bearing walls, the ratio of the shear

strength and the shear force at mechanism

shall be larger than the value in

Table 9.5.1.

Table 9.5.1 Lowest Limit of Qsu/Qmu

for Bearing Walls

Ooe / Fm Fl F2

< 0.10 1.20 1.10

0.10<, <0.15 1.25 1.10

0.15<, < 0.20 1.30 1.15

0.20<, 20.25 1.20

(2) Adequate confining reinforcement which

resists the crash of concrete shall

be provided at both ends of the bearing

wall where yield hinge is expected at mechanism. However, the bearing wall

of which 6oe/Fm is less than 0.1 and

wall width is less than 2.5m, or the

bearing wall with transverse walls is

exepted

.

(1) The ratio Qsu/Qmu required for bearing walls

In order to assure the flexural yield

failure, the shear strength Qsu is neccesarily

greater than the shear force at mechanism. The

flexural strength does not decrease so much under

repeating load, but this shear strength decreases

under repeating load in plastic range especially

under high axial force. If the value '^oe^^m

greater than 0.25 in Table 9.5.1, this bearing

wall may be the component of F3 in which the

ductility is not expected. For the bearing wall

with transverse walls, the ductility may be

113
expected and the values in Table 9.5.1 may be

decreased by 0.1, because the strength of the wall

with transverse walls is a little greater than the

calculated one and the decrease of shear strength

is small under repeating load when the transverse

walls exist at both ends of the bearing wall.

(2) Reinforcing at the ends of the bearing wall

against crash in compression

High ductility is required at the bottom of

the first story bearing wall. Since the ductility

of flexure largely depends on the limit of

compressive strain of concrete, confinement is

needed in order to protect crash after maximum

stress

.

9.5.2 Wall Girders and Foundation Beams

(1) The shear strength of the wall girders

and foundation beams of which component

type is Fl shall satisfy Eq.9.5.1.

Qsu > QL + n . Qm u (9.5.1)

where

,

Qsu:

QL

Qmu

n

shear strength of the wall

girder and foundation beam

specified in Section 9.4.2 (kg),

shear force under long term load

(kg),

shear force at mechanism (kg)

coefficient multiplied to shear

force to ensure the ductility, which

shall be larger than 1.1.

9.5.3 Reinforcement of Small Openings of

Bearing Walls

(1) In case that a small opening regulated

in the term (2) of Section 5.2 exists in

the bearing wall, additional diagonal

reinforcement and ridge reinforcement

required by Eq. 9.5.2 shall be provided

around the small opening.

(i) sectional area of additional diagonal

reinforcement at corners of the small

opening (cm^)

'td

ho + lo

2 Jl'l

Qmu

ay

(9.5.2 a)

(ii) sectional area of additional vertical

ridge reinforcement at corners of the

opening (cm^)

ho

'tv 2(l-lo)

Qmu

oy

(9.5.2 b)

( i i i ) sec t i o na 1 area of the additinal

horizontal ridge reinforcement at the corners of the opening (cm )

'th

lo h Qmu

. . (9.5.2 c)

2(h-ho) 1 ay

where

,

I

lo : ho

Qmu

ay

width of the bearing wall with

small opening (cm)

vertical distance between main

lateral supports of the bearing wall

with small opening (cm)

clear width of the small opening (cm)

clear height of the small opening (cm)

shear force of the bearing wall

at mechanism to which the value

specified in Table 9.5.1 shall be

multiplied for the component type

Fl and F2 members (kg)

material strength of the additional

diagonal and ridge reinforcement

(kg/cm^)

The Equation 9.5.2 is same as used in the

AIJ R/C Code.

Around the very small openings, about iUcm

diameter, no special reinforcement is needed in

principle

.

9.5.4 Bearing Wall-Wall Girder Joints

(1) The bearing wall-wall girder joints

which come under terms (l)-(i) through

(iii) of Section 7.8 shall de designed

against the shear force applied to the

joints at mechanism.

Application

(1) The shear force Qp in the bearing wallwall girder joints at mechanism can be

smaller one calculated by equations

below

;

Qp = Tbl+Tb2

Qwl+Qw2

(A 9.5.1)

Qp =(Twl+Tw2 - Qbi+Qb2 .) (A 9.5.2)

Dp

where

,

Tbi

^bi

: yield force of tensile reinforcement

at i-end of the wall girder( Eat-ay)

: yield force of tensile reinforcement

at i-end of the bearing wall

( lat • ay)

rshear force at i-end of thf^ wall

girder

Jll...M_

wall girder

bearing wall

Fig. A 9.5.1 Stress at Bearing Wall-Wall

Girder Joint

114
Qy-i_ : shear force at i-end of the

bearing wall

Ip : length of the bearing wall-wall

girder joint

Dp : height of the bearing wall-wall

girder joint

Oy : material strength of flexural

reinforcement

(2) The shear force used for the equations

above may be obtained from the following

way;

(i) the shear force at mechanism stipulated

in Section 9.3

,

(ii) the shear force of wall girders or bearing walls obtained in Chapter 6, (iii)

Qwl+Qw2 ll lo

2 (iv)

Qbi+Qb2

= 2 (- ^ol4

M

bl ^o2

Mb2) / (hi+h2)

(A 9.5.3)

where

,

hi

hoj_

2 ( — -Mwl + — • Mw2) / (I1+I2)

(A 9. 5. A)

flexural strength at i-end of the

wall girder at i-end

flexural strength at i-end of the

bearing wall, which may be

Z(aw • Oy) )lw span length at the left and right

clear span length at the left

and right

height of the bearing wall

clear height of the bearing wall

Fig. A 9.5.2 Notation for Design of Bearing

Wall-Wall Girder Joint

(3) The strength of bearing wall-wall girder

joints may be calculated by the following

equation

.

uQp = 0.15 • Fm • Ip • t (A 9.5.5)

where

,

Fm

Ip

t

specified strength of the RM assemblage

length of the bearing wall-wall

girder joint

thickness of the bearing wall-wall

girder joint

From the test result the average shear stress divided by prism strength uTp = uQp/(l • t) of

bearing wall-wall girder joint at maximum load was

0.18 Fcm and 0.15 Fern kg/cm.

CONCLUDING REMARKS

These design guidelines for medium rise RM

buildings were completed March 1989 by

T ECC M A R / J a p a n based upon the composite

experimental and analytical research which had

been carried out since 1984. These design guidelines are ready to be

slightly revised this year. The major points to

be revised are the specification on the wall width

ratio and the minimum amount of reinforcement

required in design method B. In a revised

edition, the wall width ratio is expected not to be specified in main body of the design

guidelines.

APPENDIX I - REFERENCES

[1 and Y. DESIGN

REINFORCED

H. Hiraishi, T. Okada, S. Okamoto

Yamazaki, "OUTLINE OF SEISMIC

GUIDELINES OF NEWLY DEVELOP

MASONRY BUILDINGS", The 2nd JTCCMAR, Keystone,

1986

[2] T. Okada, Y. Yamazaki, T. kaminosono and M. Teshigawara, "PHILOSOPHY OF STRUCTURAL DESIGN

GUIDELINES FOR MEDIUM RISE RM BUILDINGS", The

4th JTCCMAR, San Diego, 1988

Y. Yamazaki, T. Okada and S. Okamoto, "DESIGN

GUIDELINES OF MEDIUM RISE RM BUILDINGS (1)",

5th JTCCMAR, Tsukuba, 1989

T. Kubata and M. Teshigawara, "DESIGN

GUIDELINES OF MEDIUM- RISE RM BUILDINGS (2)"

5th JTCCMAR, Tsukuba, 1989

A. Matsumura, "DESIGN GUIDELINES OF MEDIUM

RISE RM BUILDINGS (3), 5th JTCCMAR, Tsukuba,

1989

[3]

[5]

APPENDIX II - CONTENTS OF DESIGN GUIDELINES OF

MEDIUM RISE RM BUILDINGS

1. General

1.1 Scope

1.2 Definicioa of Terms

1.3 Required Seismic Performance

1.4 Design CriCerior

1.5 Parallel Using of Ocher Types of Structural

System

2. Quality of Material

2.1 RM-Unit 2.2Concrete and Mortar to be used for RM

Assemblage

2.3 Strength of RM Assemblage

2.4 Materials of RM Assemblage

2.5 Concrete to be used for RC Component

2.6 Reinforcement and Wire

3. Material Strength and Allowable Stress

3.1 RM Assemblage

3.2 Concrete for RC Component

3.3 Reinforcement and Wire

4. Constants of Materials

4.1 RM Assemblage

4.2 Concrete, Morter, Reinforcement and Wire

5. Structural Planning

5.1 Size of Building

5.2 Shapes in Plan and Elevation

5.3 Bearing Walls

5.4 Small Openings of Bearing Walls

5.5 Wall Width Ratio

5.6 Wall Girders

5.7 Sub Beams

5.8 Floor Slabs

115
5.9 Cantilevered Members

5.10 Foundations and Foundation Beams

5.11 Bearing Wall-Wall Girder Joints

5.12 Others

6. Structural Calculation

6.1 Design Loads and Their Combination

6.2 Stress Analysis

6.3 Story Drift Angle

6.4 Stiffness Distribution

6.5 Uplift of Foundation

7. Allowable Stress Design

7.1 Moment and Shear Force for Design

7.2 Bearing Walls

7.2.1 Design Method

7.2.2 Allowable Flexural Strength

7.2.3 Allowable Shear Strength

7.2.4 Arrangement of Reinforcement

7.3 Wall Girders

7.3.1 Design Method

7.3.2 Allowable Flexural Strength

7.3.3 Allowable Shear Strength

7.3.4 Arrangement of Reinforcement

7.4 Sub Beam

7.5 Floor Slab

7.6 Cantilevered Members

7.7 Foundation Beams

7.8 Bearing Wall-Wall Girder Joints

7.9 Projection from Roof Slab

8. Total Flexural Resisting Moment

9. Lateral Load Carrying Capacity

9.1 Fundamental Concept

9.2 Structural Performance Coefficient D^

9.3 Calculation of Lateral Load Carrying Capacity

9.4 Ultimate Strength of Components

9.4.1 Bearing Walls

9.4.2 Wall Girders and Foundation Beams

9.5 Ductility Design of Components

9.5.1 Bearing Walls

9.5.2 Wall Girders and Foundation Beams

9.5.3 Reinforcement of Small Openings of Bearing Walls

9.5.4 Bearing Wall-Wall Girder Joints

10. Foundation

11. Joint and Anchorage of Reinforcement, Covering

Concrete Thickness and Arrangement of

Reinforcement

11.1 Joint of Reinforcement

11.2 Anchorage of Reinforcement

11.3 Covering Concrete Thickness

11.4 Arrangement of Reinforcemt

11.4.1 Bearing Walls

11.4.2 Wall Girders

11.4.3 Bearing Wall-Wall Girder Joints

11.4.4 Joint Reinforcements
U.S.-Japan Cooperative Research Program on Hybrid Control

of Seismic Response of Bridge Structures

by

Jlro TAGUCni". Toshio IWASAKP', Yoshio ADACIII^", Yasushl SASAKI"" and Kazuhlko KAWASHIMA

ABSTRACT

This paper presents current research

activities on passive and active control for

application to bridge structures conducted at the Public Works Research Institute. Japanese

-side research program for U.S.-Japan

Cooperative Research on Hybrid Control, which

Is defined as a combined approach of passive

and active control of seismic response, of

bridge structures Is also presented.

1. INTRODUCTION —

Base Isolation Is now being Increasingly

used In bridges In Japan for reducing the

structural response. Base Isolated bridge,

which Is supported by Isolators and energy

dlsslpators, causes less amount of structural

response during an earthquake, and with use of

base Isolation It Is expected to Increase seismic

safety of the bridge even during a significant

earthquake.

On the other hand, active control which can control the structural response more

effectively than the passive control Is now

being considered for application to seismic

design of structures. It is expected to bring a significant Improvement of seismic performance

of structures Including highway bridges. The

main obstacle of the active control Is the large

amount of energy required to supply to the structure. Because of such obstacle, the active

control Is now being tried to apply for

controlling the structural response against

wind.

Combination of active and passive control,

which Is defined here as hybrid control, seems attractive for solving this obstacle.

Appreciable amount of decrease of structural

response by means of passive control will make

the energy supply significantly small In the

active control to achieve the control of

structural response within a satisfactory level

during a significant earthquake. Increase of

lateral displacement of the deck Induced by

Introducing the passive control may be

Improved by the active control.

This paper presents the current research

activities conducted at the Public Works

Research Institute, Ministry of Construction, on development of design method for base-Isolated

highway bridges and a fundamental study on the

application of active control for reducing the

seismic response of a highway bridge" ^' ^

.

Japanese-side program of the cooperative

research program on hybrid control, which is now being Initiated since April 1999 at the

Public Works Research Institute, Ministry of

Construction, Is also presented.

2. CURRENT RESEARCD DEVELOPMENT FOR BASE

ISOLATED niGOWAY BRIDGES

2.1 Guideline for Design of Base-Isolated

Highway Bridges (Draft)

For studying an application of base

Isolation to seismic design of highway bridges,

a committee chaired by Professor Tsuneo

KATAYAMA, University of Tokyo, was formulated

through 1986 to 1989 at the Technology Research

Center for National Land Development. Three

programs were studied in the committee. I.e., 1) survey for Isolators and energy dlsslpators

which can be used for highway bridge, 2) study

on key points In design of base-Isolated

highway bridges, and 3) design of base-Isolated

highway bridges. As a final outcome of the

study at the committee, "Guideline for Design of

Base-Isolated Highway Bridges (draft) " was published in March 1989°'. The draft guideline Includes the following

contents:

Chapter 1 General

1.1 Related Regulations

1.2 Definition of Terms

Chapter 2 Fundamental Strategy of Base

Isolation Design

Chapter 3 Design of Isolator and Energy

Dlsslpator

3.1 General

3.2 Design Displacement of Base Isolation

Device

3.3 Equivalent Natural Frequency and

Damping Ratio of Base Isolator

3.4 Dynamic Characteristics of Base

Isolation Device

3.5 Static Characteristics of Base Isolation

Device

Chapter 4 Design of Base-Isolated Highway

Bridge

4.1 General

4.2 Seismic Coefficient Method

4.2.1 General

4.2.2 Horizontal Design Seismic

Coefficient

4.2.3 Evaluation of Natural Period of

Base-Isolated Bridge

1) Former Director General

2) Director General

3) Director, Planning and Research Administration

Department

4) Director, Earthquake Disaster Prevention

Department

5) Head, Earthquake Engineering Division,

Earthquake Disaster Prevention Department,

Public Works Research institute. Ministry of

Construction

117
4.2.4 Evaluation of Sectional Force and

Displacement due to Inertia Force

4.3 Check of Bearing Capacity for Lateral

Force

4.3.1 General

4.3.2 Seismic Coefficient Used for Check

of Bearing Capacity for Lateral

Force

4.3.3 Evaluation of Natural Period

4.3.4 Evaluation of Damping Ratio

4.3.5 Evaluation of Sectional Force and

Displacement due to Inertia Force

4.3.6 Equivalent Seismic Coefficient Used

for Check of Bearing Capacity for

Lateral Force

4.3.7 Bearing Capacity for Lateral Force

Chapter 5 Dynamic Response Analysis

5.1 Method of Dynamic Response Analysis

5.2 Modelling of Base-Isolated Bridge for

Dynamic Response Analysis

5.3 Input Motion for Dynamic Response

Analysis

5.3.1 Input Motion for Check of Design by

Means of Seismic Coefficient Method

5.3.2 Input Motion for Check of Bearing

Capacity for Lateral Force

5.4 Investigation of Safety

Chapter 6 Genferal Provision for Design of

Structural Details

6.1 General

6.2 Distance between Structures

6.3 Design Gap of Expansion Joint

6.4 Devices for Preventing FalUng-off of

Superstructure from Substructure

6.5 General Provision for Design Details of Base Isolation Device

\ppendix

I Design of Base-Isolated Highway Bridges

n Design Method of Base Isolation Device

in Example of Design Calculation of

Base-Isolated Highway Bridge

(In the Case of Ground Condition of

Group I ) IV Example of Design Calculation of

Base-Isolated Highway Bridge

(In the Case of Ground Condition of

Group n ) V Example of Structural Details

VI Example of Various Base-Isolation Device

2.2 Joint Research between PWRI and Private

Firms on Development of Bcise Isolation System

for nighway Bridge

Based on tlie study made for proposing " Guideline for Design of Base-Isolated Highway

Bridges", for aiming to develop a further

practical design method of base-Isolated

highway bridges superior from both seismic

safety and construction cost point of view, a Joint research program between PWRI and 29 private firms on "Development of Base-Isolation

Systems for Highway Bridges" was initiated since

1989. The research is scheduled to continue for

three years until 1992. Scopes of the Joint research contain the

following research/development programs :

1) Development of isolator and energy

disslpator suitable for highway bridge with

utilizing new materials and technologies

2) Development of expansion Joint and

falUng-off prevention device suitable for

base-Isolated bridge

3) Development of design method for

base-Isolated bridge

4) Application of base-isolation for highway

bridge

In the project 1), It is aimed to develop

less expensive Isolator and energy disslpator

which are superior for long-term use as well as function. In the project 4), application of base

Isolation for super-long multiple-continuous

highway bridge with total length over 1 km and

for seismic retrofitting of vulnerable existing

highway bridges is being Investigated.

The 29 private firms Joined In the Joint

research program consist of material makers

such as rubber and automobile tire makers,

bearing supports fabricators, consulting

engineering firms, steel superstructure

fabricating companies, and general

constructors. Table-1 shows the research subjects and

contributions.

2.3 Construction of Base-Isolated Uighway

Bridges

Seven base-isolated highway bridges as shown in Table-2 are currently being

constructed by the support of the Ministry of

Construction for the purpose of Incorporating

the base Isolation design into the practical

use. After completion of the construction, a series of tests Including a push-pull test, forced excitation test with use of an eccentric-mass shaker and strong motion

observation Is scheduled to be made for

studying the dynamic characteristics of the

bridges.

Among the seven bridges presented in the

Table-2, the design has been already completed

for Nagakl-gawa Bridge, Marukl Bridge,

Karasuyaraa No.l Bridge and Mlyakawa Bridge,

and construction of the substructures Is now

under way.

3. A FUNDAMENTAL RESEARCn ON APPLICATION OF

ACTIVE CONTROL TO BlCnWAY BRIDGE

3.1 Seismic Response Control by Means of Active

Mass System

To study an application of active control

for reducing seismic response of bridge, a simple analysis idealizing the bridge by a one degree of freedom system Incorporating an

active mass on the deck was made*'. Fig. 1 shows

an analytical model with an active mass. Equations of motion of the bridge with the

active mass system can be written as

118
m 8 0

1

0' is,

0 ma 0 0 X J

k a*- K d - k d' (X. ('•' '~"\ +

1 - k d k d 1I f ^ ul

(1)

where,

IDs, , ks : mass, damping coefficient and

stiffness of the bridge

s, Xs, Xs : relative displacement, velocity

and acceleration of the bridge

fa : external force applied to the bridge

ffld, kd : mass and stiffness of the active mass system

1, Xd. Xd : displacement, velocity and

acceleration of active mass

relative to the bridge

fa : external force applied to the active mass

u : control force of active mass system

The control force Is assumed to be

determined by a linear combination of

displacement and velocity of the bridge and the

active mass as

u = [ g' ;]!xsi+ [G G d*^

(2) In which, Gs° and Gd° represent control gain

vector of velocity and displacement of the

bridge, respectively, and Ga*^ and Gd"^ represent

control gain vector of those of the active mass, respectively. Substituting Eq.(2) Into Eq.(l), one can obtain the equations of motion of the

total system as

It Is an Important role to determine

appropriate control gain vector G In order to adequately control structural response using

active mass system. Various methods to determine control gains has been studied and

proposed In the literature. In this study, the

following two control methods are considered.

1) Optimal Feedback Control Method

In the optimal feedback control method,

control gains have to be determined so that the

objective function J be minimized base on an optimal control theory..

J =

1 ( X

"^Q X + R u d t (5)

In which matrix Q and R denote weighting

factors on state variables and control force,

respectively.

Control gain vector G can be obtained by

matrix P which satisfies Algebraic Rlccatl

Equation as

A"^P + PA + PBR (6) (7)

2) Modal Control Method

Modal control method gives a control force

as a linear combination of modal responses, which are generally called as generalized

coordinate. The control force can be obtained

as

u = E k r J (8)

In which Kj and rj represent control constants

and J-th generalized coordinate, respectively.

Control constants are determined using a method developed by Porter et.al as X=AX+Bu+D (3) where

A_ =

M

Xa Xd Xa Xd} ~^

-I -1 -ms ma 0 0)^

— 1 —I — mats m a f a 0 0 M.= fma 0 1 [0 m dj C.= r C a

0^ [o 0 K.= fka+kd -k

- k d k

I : unit matrix

Introducing state variables X, the control

force by Eq. (2) can be rewritten as

u = G_X.

In which

(4)

G d)

TT (p ,- A j) m

(9)

In which

P I

complex eigenvalue of j-th mode of

uncontrolled system

complex eigenvalue to be assigned of

J-th mode of controlled system

j-th element of a vector obtained

from complex modal vector, and

matrices A and B

3.2 Bridge Analyzed and Analytical Conditions

Fig. 2 shows a bridge analyzed which Is of

three span continuous girder bridge with deck

length of 96 m. Response In longitudinal

direction Is considered for the analysis.

Fundamental natural frequency is evaluated as

0.99 Hz. Damping ratio Inherent of the bridge Is assumed as 0 for aiming to Isolate the effect of

119
the active mass system.

Since the selection of the proper mass Is one of the Important design parameters In the

active mass system, 3 cases with a mass ratio,

which Is defined as a ratio of the active mass to the mass of the superstructure ( 1,835 tf ), of

1/16. 1/160 and 1/1666 are considered In the

analysis. Spring coefficient ka of the active mass system Is assumed so that the natural

frequency of the active mass system becomes 1/4

of the fundamental natural frequency of the

bridge ( 6.99Hz ). Control gain vector is determined in the

following ways:

1) Optimal Feedback Control Method

Weighting matrix Q Is assumed to be as

0 0 0

0 Q d 0 0

d 0 0 Q a 0

0 0 0 Q

(16)

in which, Qa" and Q^" denote weighting factors

for relative displacement and velocity of the

bridge, respectively, and Qd" and Qd'' represent

those of active mass. Because bending moment

developed at the pier which Is the most important parameter for seismic design of

substructure is generally proportional to relative displacement of the bridge, the

control force is assumed to be determined only

by Qsd (I.e., Qs"=Qd '~'=Qd "=6) in this study. The

weighting factor of control force R is fixed as

1.0, and three cases with Q^" of 16°, 16'' and 18'° are considered. Table-3(a) shows the control

gain vectors thus determined.

2) Modal Control Method

Control gain vector is determined so that

damping ratio of the total system becomes 6.1, 0.2 and 6.3 without changing natural frequency

of the bridge. Natural frequency of the bridge

was not changed so that effect of the input

ground motion on seismic response be the same

with the analysis assuming the optimal feedback

control. Tab)e-3(b) shows the control gain vectors thus determined.

Two ground motion acceleration records

obtained at ground surface near the Kalhoku

Bridge during the Miyagi-ken-oki Earthquake (M

7.4) in 1978 (referred as Kalhoku Record) and on

the llatlro-Gata during the Nlhonkal-chubu

Earthquake (M 7.7) In 1983 (referred as Hachlro-gata Record) are used for the input

motion. Peak acceleration of the two records

was assumed as the same with the ones actually

measured.

3.3 Effect of Active Mass System Assuming

Optimal Feedback Control Method

Figs. 3 and 4 show the response displacement

and acceleration of the bridge assuming the

optimal feedback control method when subjected

to the Kalhoku Record. The mass ratio ma/ms of

1/106, and weighting factor Q^" of 16° and 10'° are assumed. According to these results, when

Qa" Is assumed as 10°, sinusoidal response is seen In both displacement and acceleration,

which implies that an apparent damping ratio of

the system is quite small. Predominant

frequency (0.99Hz) of the response corresponds

to the natural frequency of the system

consisted of the bridge and the active mass. Peak response displacement of the deck is 6.8cm. On the other hand, when Qa" is taken as

16'° both displacement and acceleration

responses become quite small as compared with

the case assuming Q^" as 10°. Acceleration

response has a similar shape with the Input

ground acceleration. This Implies that apparent

damping ratio seems very large.

Effect of active control system is studied

In terms of peak response. Flg.5 shows how peak

response displacement and acceleration of the

bridge vary in accordance with the weighting

factor Qa" and the mass ratio ma/ma. Peak

response displacement decreases monotonously

in accordance with increase of Q^" for both

Kalhoku and Hachiro-gata Records. However,

peak response acceleration does not decrease

monotonously, and has its peak at Qa" = 10*^ for

the Kalhoku Record, it should be noted that In some cases the response acceleration does not decrease monotonously even if control gain

vector increases. This Is attributed to the

stroke of the active mass, as will be discussed

later.

Fig. 6 shows stroke of the active mass. For

the Kalhoku Record with the control of Q," = 10'°, peak control force becomes 199 tf which

corresponds to 19% of the dead weight of the

superstructure (l,835tf). The stroke of the

active mass to get the peak control force of 199 tf significantly depends on the mass ratio

ma/ma. When r^ = 1/1066 the peak stroke of the

active mass becomes extremely large, i.e., 37.4

m. It Is impossible to ensure such large stroke

of ±37.4 m in the bridge with the span length of

38ra. On the other hand, when rm = 1/10 the peak

stroke of the active mass becomes 0.42 m, which

may be In the range of a practical use. However,

although the stroke of the active mass is within

a tolerable range, the mass with 16% of the dead

weight of the superstructure Is excessive for a practical use. Therefore, it Is quite important

to design the system so that the mass and the

stroke are In the tolerable range.

3.4 Effect of Active Mass Assuming Modal

Control Method

FIg.7 shows peak response displacement and

acceleration computed assuming the modal

control method. Peak responses of the bridge

decrease monotonously in accordance with Increase of the damping ratio of the system hs. This Is quite natural because damping ratio of

the system Is assumed to increase by means of

120
the active mass system. FIg.8 shows the control

force and the stroke of the active mass. As was

the case of the optimal feedback control, the

control force Is not affected by the mass ratio,

because it is compensated by the stroke of the

active mass.

4. U.S. -JAPAN COOPERATIVE RESEARCH PROGRAM ON

OYBRID CONTROL OF STRUCTURAL RESPONSE

Although the active control Is an attractive

approach, energy required to supply for

controlling the seismic response of bridge structures within a desirable level Is excessive. This obstacle seems quite difficult

to solve with use of the the currently existing

technology, In particular In Japan where occurrence of an earthquake with magnitude over 8 has to be taken Into account In seismic

design. Hence, It Is considered difficult to apply active control system for controlling

seismic response of bridges In near future.

Therefore, at this stage a combined

approach of passive and active control, which

Is defined here as hybrid control, seems more

superior for controlling seismic response of

bridges.

The advantages of the hybrid control can be

expected as

1) By Introducing active control to a base-Isolated bridge, deck response

displacement, which sometimes becomes excessive

for avoiding to use expansion joint with large

gap, can be reduced. Thus, the hybrid control

extends the applications of base Isolation.

2) Control energy can be significantly

reduced as compared with the energy required

In active control.

3) Dependence on active control can be

decreased, which. In turn, decreases

inconvenience caused by ralss-worklng or malfunction of the control system.

Base on such considerations, the hybrid

control seems more superior than the active

control for aiming to reduce structural

response of bridges during a significant

earthquake. Therefore, a 5-year research

program on hybrid control of seismic response

was Initiated In the Public Works Research

Institute from April 1998. The objectives of this research program are to study an application of

the hybrid control for bridge structures (reter

to Flg.9) as

1) Control of super long-span continuous

bridges

2) Control of bridges on relatively soft

ground

3) Improvement of seismic performance of

bridges on the Important urban route

For such targets, the following research

Items are considered for study :

1) Development of passive control element

suitable for hybrid control system

• Development of variable damper

• Development of damper with fall-safe function

2) Development of active control system

suitable for hybrid control

3) Development of optimal hybrid control

system by combining 1) and 2)

This research program Is to be executed as

a U.S. -Japan cooperative research program

through the activities of the Panel on Wind and

Seismic Effects of UJNR.

REFERENCES

1) Kawashlma, K. : Development and Future Scope

of Seismic Isolation of Structures - A Review -, Proc. of Japan Society of Civil Engineering,

Structural Eng./Earthquake Eng., Vol.398/I-lQ, 1988

2) Kawashlma, K. : Research Needs for

Development of Seismic Design. Seismic Isolation

and Active Control Technology, Proc. JSCE,

Special Issue for Disaster Prevention, 1989

3) Kawashlma, K. et al. : Shaking Table Test on Dynamic Response of Base-Isolated Bridges,

Civil Engineering Journal, Vol. 36-10, 1988

4) Japan-New Zealand Workshop on Base Isolation

of Highway Bridges, Head Office of Ministry of

Works and Development, Wellington, New

Zealand. 1987

5) Guideline for Design of Base-Isolated Highway

Bridges, Final Report of A Committee for Study

of Application of Base Isolation to Seismic

Design of Highway Bridges (Chairman : Professor

T. Katayama), Technology Research Center for

National Land Development, March 1989

6) Kawashlma, K., Hasegawa, K. and Yoshlda, T. : Active Control of Seismic Response of Structure

by Means of Mass Damper, Civil Engineering

Journal, Vol. 31-5, 1989

121
^d

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122
Tabic 2 Construction Project of Basc-Isolatcd Highway Bridge

Owner Bridge Name Type of Superstructure Length ( m)

Holdcaido Developing Bureau Onnetoh Bridge Steel Girder

Tohoku-RegionaJ Bureau, MOC Nagaki-gawa Bridge // 97

Kantoh-Regional Bureau, MOC Not Yet Selected Not Yet Decided Not yet Determined

Chubu-Regional Bureau, MOC // // //

I wate-ken Maruki Bridge Prestressed Concrete 92

Tochigi-ken Karasuyama Bridge // 250

Shizuoka-ken Miyagawa Bridge Steel Girder 110

Table 3 Control Gain

(a) Optimal Feedback Control

Case Gain m d/m G; [N s/m] G'^ [N-s/m] [N/m] [N/m]

110

10' 160 -5.34 22 -26 10

«

1,560 -52.2 2.100 -2.560

10 '° 9,630 -322 79,800 -97,300

1 100

10* 160 -0.54 22 4 10' 1,570 -5.24 2,120 378

10'" 9,640 -36.2 79,900 14,300

1 1000

10' 161 -0.05 22 -1

10

«

1,570 -0.53 2.120 -66

10 9,640 -3.22 80,000 -2.480

(b) Modal Control

Case Gain

m d/ m 1 h, G; [N-s/m] G', [N-s/m] G^ [N/m] G^ [N/m]

110

0.1 1.210 -10.5 402 763

0.2 2,420 -48.5 804 750

0.3 3,640 -56.6 1,210 653

_L_

100

0.1 1,210 -4.05 402 139

0.2 2,420 -4.85 S04 133

0.3 3,630 -5.66 1,210 127

1 0.1 1,210 -0.40 402 13.9

1000 0.2 2,420 -0.49 8(W 13.2

0.3 3,620 -0.57 UIO 12.5

123
Ground Motions

Fig. I Analytical Model of Onc-Dc|^rcc-or-Free(loin System

with Active Mass Damper

0 ,5 90 0, 5

F

30 30 30

2 F\ 1

CM

CM

CM

CM

CMCM

CM

CM r c.

11 0.4 2.5 7.5 0.6

»—

1

4.5 2 4.5

eg

UNIT fml

Highway Bridge Analyzed

124
500

2501-

0 250

•500

I I I I I I I I 1 I I I I I I I I i I I I I I I I 1 I I 1 I I I I I I I I ifSFD

0 5 10 15 20 25 30 35

(a) Acceleration

100

50

0 -50

•100

I I I 1 I I I I I I I I I I I I I :

I I I I I I I I I I 1 I I I I I I I I I I i(SECl

0 5 10 15 20 25 30 35

(b) Displacement

I I t I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I r cpr

I

0 5 10 15 20 25 30 35

(c) Input Acceleration

Fig. 3 Computed Response by Optimal Feedback Control

with Qo*" = 10° (mc/mo = 1/100)

500

250

0 -250

-500

I I I I I I' I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I irsFD

0 5 10 15 20 25 30 35

(a) Acceleration

100

50

0 -50

100

I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I Irecp)

0 5 10 15 20 25 30 35

(b) Displacement

#1

400

200

0 •200

•400L

I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I r cpp ] 0 5 10 15 20 25 30 35

(c) Input Acceleration

Computed Response by Optimal Feedback Control

witii Qo^ = 10'° (ma/ma =1/100)

125
10

£

8

o

i3 6 a(/) 5

0) 4 c/o C

O

a

^ 2

Oh n

7m =1/10

7m = 1/100 7m = 1/1000

Hachiro-Record

Kaihoku-Record

10* Coefficient Q'i

10'

ccco

u

0) <

c

oc

o

400

300

200 -

100

7m

7m

1/10 1/100

—7m =1/1000

iro-Record

Kaihoku-Record

10* 10' Coefficient Q'l

10'

Fig. 5 Control of RcsDonse by Means

or Optimal Feedback Control

E

aQ

>

CJ <

O

(Zl

100

10

1 -

0.1

1 "T 7m= 1/10

1

7m = 1/100 7m = 1/1000--^

\ Hachiro

) -Record

Kaihoku r^'^^^^'

- -Record^^-'V'^ /

1 1 1 10* 10' 10'°

Fig. 6

Coefficient Qs

Stroke Required by Optimal

Feedback Control

co

ctn<V K

^ a

200

S. 2

'^100

Oh <

Kaihoku-Record

Hachiro-Record

7m=l/10 —7m= 1/100 7m = 1/1000

_J L 0.1 0.2 0.3 Damping Ratio /i.

n 1 r Hachiro-Record

Kaihoku-Record

.7m= 1/10 7m= 1/100 7m = 1/1000

0.1 0.2 0.3 Damping Ratio /i,

I^'ig.T Control of Response by Means

of Modal Control Method

100

^3

(J V ^ £ o a

O

c/3 S

10

0.1

Kaihoku

-Record

0.2 0.3 Damping Ratio

Fig. 8 Stroke Required by Modal

Control Method

126
0)coa

\ \

127

Theme IV

Earthquake Engineering

Lateral Resistance of a Pile

BY

Michio Okahara^ , Shoj i Takagi^ and Keiji TaguchiABSTRACT

In order to investigate static

horizontal resistance of piles, numerous horizontal loading tests were performed and the results

analyzed using statistical

techniques. Through analysis of

these horizontal loading tests,

we examined factors influencing

amounts of yield displacement and

evaluated both amounts of allowable

displacement and coefficients of

ground reaction in a horizontal

direction, assuming that yield

displacement is defined as the

amount of displacement at the point

where residual displacement

rapidly increases with respect to

horizontal displacement.

KEY WORDS: Pile, horizontal loading

test data, yield displacement,

allowable displacement, horizontal

ground reaction coefficient

In Secifications of Substructures

in "Specification for Highway

Bridges" Part IV (hereafter referred

to as Guide), the linear elastic

ground reaction method is adopted,

which assumes that the ground

reaction is proportional to pile

displacement and which sets

coefficients of ground reaction in

a horizontal direction and

allowable amounts of horizontal

displacement for piles of relatively

small diameters. However, since both

the size of pile foundations

and varieties have been increasing

with bridge scale, a more

rational approach must be

established when designing

bridges. In this report, we have

investigated horizontal resistance

properties of piles by analyzing

data obtained from loading tests

by means of statistical methods.

2. SUMMARY OF HORIZONTAL PILE

LOADING TEST DATA

1. INTRODUCTION

It has for many years been

necessary to establish a method

of predicting pile behavior

under conditions where horizontal

force and moment act on a pile.

In the case of static behavior

alone, this subject is complicated

by the dependency of pile behavior

on various factors each such as non-uniformity and non-linear

characteristics of the ground,

the type of pile, and the method of

pile construction.

Approaches to predicting

horizontal behavior of a pile can

be generally classified mainly

into the following three categories

regarding the concept of horizontal

ground reaction:

1 ) The ultimate ground reaction

method

2) The (linear or non-linear ) elastic ground reaction method

3 ) The elastic-plastic ground

reaction method

Horizontal loading test data were collected for the 415 cases to be

analyzed. The primary data

( Table 1 ) available for the

following statistical analyis were selected from among these data,

which were important to our investigation. Figures 1 and 2 show

the frequency distributions of pile

diameter and length, respectively.

In the following sections: "3.

Studies of elastic limitation on

the horizontal behavior" and "4.

Studies of the coefficient of

horizontal ground reaction"

,

statistical methods were applied to

the secondary data groups selected

Foundation

Structure and

Public Works

Ministry of

^Engineer Head,

Engineering Division

Bridge Department,

Reseach Institute,

Construction

^Senior Research Engineer,

Foundation Engineering Division, Ditto

^Research Engineer, Foundation

Engineering Division, Ditto

131
so that validity was retained for

respective purposes and methods

(Table 2 and 3 )

.

3. INVESTIGATION ON LIMIT OF

HOR I ZONTAL ELAST 1 C BEHAV 1 OR

3 . 1 Yield displacement and its definition

When the horizontal force and

moment act on the pile

foundation, its behavior in a horizontal direction is usually

non-linear. This means that soils

are not perfectly elastic and a region where plastic deformation

occurs gradually grows larger with

increasing loads. In the Guide,

the non-linear behavior of pile

foundation is treated as the

"apparent" linear behavior,

utilizing the coefficent of

horizontal subgrade reaction ( Kh ) obtained in a linear condition.

This is because elastic

design within a region of

allowable horizontal displacement

should enable soil resistance

to an immense and repetitive load,

such as seismic force. We have

here, from a similar point of view,

investigated with respect to

the displacement necessary to

attain not only mechanical

stability of a pile, but also

to withstand an increase of

harmful residual displacement.

Figure 3 shows the relationship

between residual displacement of a pile head after removing the load

and its strain ( which is defined

as the ratio of the amount of pile

head displacement to the pile

head diameter before removing the

load). From the figure, which is a typical example, one can see that

residual displacement rapidly

increases beyond a certain amount

of pile head strain. It is thus

suggested that large plastic

deformation of soils could not

only be prevented but also its

stability as a foundation could

be preserved if the pile head

displacement were designed to be

below the point of inflection

where the amount of residual

displacement begins to increase

rapidly. The following

investigations were carried out

by defining the amount of pile

head displacement corresponding

to that of the inflection point for

residual displacement, to be the

yield displacement.

3.2 Results Of Yield

Displacement Analysis

The amount of yield displacement

of a pile derived from data is

affected by various factors such as pile size, type, method of

construction, and hardness or softness of the soil. Yield

displacement was evaluated for

respective data, and correlations

examined between the following

indices: yield displacement ( Sy )

,

ratio of the yield displacement to

the pile diameter ( Sy/D ) , type of

pile and modulus of the ground

deformation ( Eg o ) • Figvires 4 and 5 present

relationships between Sy vs. Eb o and Sy/D vs. £50- Here, the modulus

of the ground deformation E5 0 was

N up to depth 1/ 0 .

( /3 = ( K„ D ) / ( 4EI) ) Since the results show great

variation, there does not seem

to be any index governing variation

of yield displacement as seen in

the figures. Therefore, relative

frequency of Sy/D, and the

average values and coefficient of

variation were obtained based on statistics. By assuming that the

frequency distribution of Sy/D

is a normal logarithmic

distribution, probability of

non-excess for three cases (10%, 15,9%

and 30% ) were checked

.

Results of such statistical

treatments presented in Figure 6 and Table 4 are summarized as follows:

( 1 ) The coefficients of variation for

respective types of piles obtained

from Figure 4 fall in the region

of 76%-105%, showing a wide

distribution. This observation

may be due to unknown

characteristics of soils

and artificial causes at the

time of pile construction and

loading tests.

( 2 ) Statistical values obtained for

cast in-situ piles seem to be

different from those for other

types of piles. This is because

the quality of cast in-situ piles

varies greatly with the quality of

construction, and because soils

around a pile are likely to

soften excavation works, while

distortion and cracking of a pile

132
resulting from loading tests

change its section rigidity

and a pseudo-yield point is

therefore likely to be formed.

(3) The average value of Sy/D is

evaluated as 3.8% for the

total distribution of all

piles, and those for three

probabilities of non-excess are

in the range of 0.9 to 1.7%.

Considering the probability of non-excess of 10 to 15.9%, a figure approximately 1% of the

pile diameter is a reasonable

index, including the amount

of yield displacement.

3 . 3 Displacement Allowable In

Designing Pile Foundation

Since a foundation is an underground structure, it is

necessary to set an appropriate

model of calculation, critical

terms and allowable values with

respect to its design. In pile

design for highway bridges in

Japan, ground with non- linear

physical characteristics is treated

as having linear characteristics

while, horizontal behavior of a pile has been checked by using

the linear method of elastic

ground reaction, and horizontal

stability of a pile foundation

has been examined by measuring

amounts of horizontal

displacement at the design ground

level. The concept of yield

displacement described in the

previous section is derived from

the concept that the amount of

displacement corresponds to

the limit of horizontal elastic

behavior of piles and soils. As

long as the amount of displacement is

below that of yield displacement,

the assumption that soils are elastic is reasonable and no significant errors should occur

when designing.

Therefore, the yield displacement

defined above can be used as the

allowable displacement, which

prevents formation of harmful

residual displacement and

restrains horizontal behavior

of a pile, keeping it within

its elastic regime. These results

suggest that the horizontal

displacement allowable in designing

should be rationally estimated by

a certain ratio, approximately

1%, of the pile diameter.

4. INVESTIGATION OF COEFFICIENTS OF

HORIZONTAL GROUND REACTION

4.1 Subjects

The ground reaction coefficient has

complicated characteristics: strain

dependability, dependability on load width, dependability on load

time. Since calculated values of the

ground reaction coefficient vary

with strain level, strain level

must be evaluated when designing.

The fact that since the piles

foundation is elastic, the depth of

foundation able to resist loads is

well known to change with pile

diameter and rigidity, and with

softness or hardness of the

ground must also be considered.

Moreover, since types of foundation

vary according to purpose, the

interation between the pile and

the soil must be investigated. We

have investigated these points

using data obtained by

horizontal loading tests of piles

as follows.

4 . 2 Methods and Results

The strain level due to loads

acting on the pile head was assumed to be 1% of the pile

diameter in order to investigate an equation for estimating ground

reaction coefficients in a horizontal

direction. This is the standard

amount of displacement so that the

horizontal reaction coefficient is

calculated within the region of

elasticity in the behavior of pile

and soil. In order to examine

possible timbling of the pile and

characteristics of the ground, the

loading area of pile(AH) is given

by Eq. 4. Eq.l is the equation

used to estimate the ground

reaction coefficient:

Kh = Kho (Bh/ 30 ) -^'^ (1)

Here, KH:the coefficient of

horizontal ground reaction

( Kg f /cm^' ) Kho : the coef ficent of gruond

reaction in a horizontal

direction (Kg/cm^)

equivalent to the value

133
determined from a flat plate

loading test conducted by means of a rigid disk of

30cm diameter and obtained

by the equation:

Kho = Eo/30 (2)

Bh : the equivalent loading width

of a pile ( cm ) perpendicular to

the direction of loading and

obtained by the equation:

Bh =^ (3)

Ah : the loading area of a

( cm^ ) perpendicular to

direction of loading

obtained by the equation:

Ah = DX 1/ (3 (4)

D: the diameter of pile

1/|3 : the depth of

related to

resistance ( cm

)

the property value of a pile

pile

the

and

; cm

)

foundation

horizontal

/3

EI:

'Kh D/(4EI)

the bending

pile ( Kg cm^ )

of

( cm' ' rigidity of

Using the secondary data group

shovm in Table 3, reversecalculated values of Kh were

obtained for the 1% level of pile

head strain from individual

loading tests. These values were

then compared with those

estimated using Eq.l from results

of the basic survey carried

out at the same positions. Here,

the reverse-calculated values of

Kh were obtained from amounts of

displacement at the ground

surface of pile using the

Hayashi-Chang Equation.

Fig . 7 compares the reversecalculated values of Kh with

those of Kh calculated by Eq.l.

The broken lines denote turning

lines related to the least-square

method, and the solid lines show

the regions of 1:1, 2:1, and 1:2,

respectively. Fig. 8 shows

the normalized probability

distribution form ratios of the

reverse-calculated Kh to the Kh

calculated by Eq.l, assuming a normal logarithmic distribution.

The following observations were

obtained from these results:

( 1 ) A good correlation does not

appear to exist between the

reverse-calculated Kh and the

estimated Kh , revealing

considerably scattered data.

( 2 ) Estimated Kh tends to be

slightly larger than the reverse- calculated Kh . (3) The method, in which 1-% of the

pile diameter is adopted as the

standard displacement for

calculation of the horizontal

ground reacion coefficient, shows

applicability to various types

of piles, while the use of Eq.l

was found to be generally

suitable for estimating the

ground reaction coefficient

in a horizontal direction.

5. CONCLUSION

In this study we have examined

resistance characteristics of piles

in a horizontal direction by using

the elastic ground reaction method

in the pile design. Results suggest

that yield displacements in loading

tests of piles are approximately 1%

of the pile diameter, and that the

yield displacement seems to correspond

with the diameter of the pile.

Moreover, from these results, we have

proposed an equation able to

estimate both allowable

displacement and coefficients of

ground reaction in a horizontal

direction on the design ground

level; error widths of data

obtained were relatively large

because loading tests were carried

out for soils whose mechanical

properties are not well known.

Although this paper has

dealt with static loading

tests, it is also important

to study resistance characteristics

of dynamic loads due to actual

seismic force. It will be necessary

to carry out further investigations

regarding such subjects,

including characteristics of

dynamic pile resistance in a horizontal direction.

REFERENCE

1 ) Japan Road Association;

Specifications of Substructures

in the "Specifications for

Highway Bridges" part FV , May

1980, February 1990.

2) Uto, Maeda, Gose; "An Investigation

134
of the Standard Displacement of a Pile in a Horizontal Direction",

Bridge and Foundation, Sept.

1987.

3) Okahara, Takagi, Nakatani,

Taguchi, Sakamoto; "Survey

of Horizontal Resistance

Properties of Piles by Loading

Test Data", Public Works

Research Institute Data, 2721,

Jan.

4) Okahara, Takagi, Nakatani,

Taguchi, Sakamoto; An

Investigation of the Standard

Displacement of Pile by Data from

Horizontal Loading Tests", No.l

and 2, Proceedings of the 24th

Conference on Soil Mechanics,

1989.

135
^

PHC piles

I I PC piles

Cast in-situ piles

Steel pipe piles

30 30 40 50 60 70 80 90 100 110 120 130 140 150 170 190 210

S \ \ \ S S S \ \ \ S 40 50 60 70 80 90 100 110 120 130 140 150 170 190 210 230

Pile diameter D (cm)

Figure 1. Frequency distribution of pile diameter

60

5 10 15 20 25 30 35 40 45 50

^ ^ ^ ^ ^ \ ^ ^ ^ ^ 10 15 20 25 30 35 40 45 50 55 pile length £ (m)

Figure 2. Frequency distribution of pile length

136
2 -

0

Experimental

Cast in-situ pile

^z;- 1 0 0 0 mm, £ = 1 1 m

S r y/D= 0.43 %

S r y = 4 . 3 mm

—I 1 I ' 0.5 1.0 I. 5 Sry/

Pile head strain (S/D) (%)

Figure 3. An example showing a sudden increase point of

residual displacement obtained from loading test data

ooo

CI tn D

CZ3

^^e>..:°-?>.*

o "^D + o fiat's * D ; _ a Ok n cm o^'fo ^

«

f8 o

*°* o o

o Q

pa

o & oo

10

Figure 4. Relation of Sy ~

137

100 1000

E 50 ( k g f /c m2 ) Eg 0 for all piles
0

e

CD

e

! °

+A ai 1 D

: 0

a Q

i g

* 0

e

; 0

D

e

' • »

OA g^ff*

°

? B A 5^ DO D Cj H -o t»-S- D

A Q O D

. g ^

o

D-0

o e e

o--e o e D 0 o

e

%° o °

O Jo

5 + + D • -oo OA o

o

a

D

e

10 100 1000 E 5 0 ( k g f / c m 2 ) Figure 5. Relation of Sy/D~ o for all piles

«— Total

Steel pipe piles

Sy/D (%)

Figure 6. Frequency distribution of Sy/D

138
Estimated value of K« (tf/m*)

Figure 7. Comparison of the reverse calculation

Kh with the estimated Kh

0. 100

0. 075

0. 050

0. 025

0. 000

(average) 1.548

(standard deviation) 1.480

(coefficient of variation) 0.96

-4.0 -3.2 -2.4 -1.6 -O.B 0.0 0.6 '..6 2.4 3.2 ln( reverse calculated K„ /estimated K„ )

4.0

Figure 8 Relative distribution of ln{ reverse calculated

Kh /estimated Kh ) 139
Table 1. The number of data collected from loading tests and the number in

the primary data group

Collected data Primary

UaXia

First selection terms

Steel piles 237 137 Q Those whose standard penetration

test data was insufficient

Cast in-situ

piles

100 30 (2) Those consisting of unknovm

piles components

PC piles 48 40 d) Those use special types of

pile and construction method.

PHC piles 30 24

@ Those whose curve of Weibull

distribution is larger than

the Pmax obtciind from loading

tests: Py>Pmax

Total 415 231

Those whose data of load~

displacement do not follow the

Weibull distribution curve.

Pmax : Maximum loading load py : Yield load estimated

Table 2. The number in the secondary data group and selection terms ( Allowable

)

Primary Secondary Second selection terms

data data

Steel piles 137 117 (1) Those whose height of loading

load from the ground surface is

Cast in-situ 30 16 larger than the pile diameter

piles H>D

0Those whose Smax obtained

from loading tests does not

PC piles 40 31 reach 1% of the pile diameter

Smaa{<0.01D

©Those whose circle number of

PHC piles 24 24 measurement for load~ displacement

under the loading test is less

than 2 @ Those in which 80% of the Pu

shown as a Weibull distribution

curve is larger than the Pmax

Total 231 188 obtained from loading tests

0. 80Pu>Pmax

Smax : Maximum horizontal displacement Pu : Ultimate load estimated

Pmcix : Maximum loading load

140
Table 3 . The secondary data and its selection terms ( Kh )

Primary

data

Secondary

data

Second section terms

Steel piles 137 117 (D Those whose estimated

yield load is larger than

the Pmax obtained from

loading tests and whose

Smax is less than 1% of

the pile diameter.

(2) Pile groups , those

whose pile head is fixed

and whose heights are not

known.

1 %

Cast in-situ

piles

30 16

PC piles 40 31

PHC piles • 24 24

Total 231 188

Smax : Maximvun horizontal displacement Pmax : Maximum loading load

Table 4. Results obtained from statistical treatments of Sy/D

Steel pipe

piles

Cast

in-situ

piles

PC piles PHC piles total

Sy/D

Range ( % ) 0.3 ~ 22.4 0.3 ~4.0 0.7~ 11.3 1. 1~ 14.9 0.3~ 22.4

Average ( %

)

4.2 1.3 3.6 3.8 3.8

Standard

deviation( % )

4.4 1.1 2.8 2.9 3.9

Variational

coef f icient( %

)

105 85 78 76 103

Probability of non-excess

10.0% 1.0 0.4 1.1 1.3 0.9

15.9% 1.2 0.5 1.4 1.5 1.2

30% 1.9 0.7 2.0 2.1 1.7

141
Major Seismic Building Upgrades at the

Lawrence Livermore National Laboratory

F.J. Tokarzi' G.E.

ABSTRACT

The Lawrence Livermore National

Laboratory is a U.S. Department of

Energy research complex located near

San Francisco and Oakland,

California. The Laboratory,

established in 1952, is located in a highly seismic region which is

surrounded by both major (San

Andreas) and local fault systems.

In fact, in 1980 the Livermore

Earthquake (M-5.8) caused

considerable damage at the

Laboratory. As a result, the

Laboratory has spent some $25

million (1990 U.S. dollars) in a seismic upgrade program to enhance

life safety of each employee, and to

minimize potential damage against

future earthquakes. This paper

focuses on the major seismic

upgrades (retrofits) of three

buildings: (1) a two-story physics

office building, (2) a five-story

computer office building and (3) a seven-story administration building.

Also, included herein are brief

descriptions of the Laboratory,

seismic setting, impact of the 1980

Livermore Earthquake on Laboratory

facilities, and a discussion of the

seismic design criteria employed

over the years, with particular

emphasis on criteria used since the

1980 Livermore Earthquake.

by

Freeland^' and J.R. Hill^

1. INTRODUCTION

1 . 1 Description of Laboratory

Lawrence Livermore National

Laboratory (LLNL) was established in

1952. The Laboratory is operated by

the University of California for the

U.S. Department of Energy (DOE).

The Laboratory site is approximately

one square mile (640 acres) [2.59

km^] in size and contains municipal

services similar to those normally

found in any small city; police and

fire department, a medical

department, and

water/sewer/electrical /natural gas

distribution systems. Presently,

personnel at the site number around

11,000. There are approximately 200

permanent buildings that enclose

about 3.9 million sq. ft. [3.62x10^

m^ ] (gross) of floor space, and over

Division Leader, Plant

Engineering Department,

Lawrence Livermore National

Laboratory, POB 808 L-654,

Livermore, CA 94550

Civil and Structural Section

Leader, Plant Engineering,

Lawrence Livermore National

Laboratory POB 808 L-654,

Livermore, CA 94550

Manager, Natural flazards

Mitigation Program, U.S,

Department of Energy, 19901

Germantown Road, Gerinantown, MU

20874

142
1,000 trailers on site (single,

stand alone units or joined together

to form one or two story complexes)

.

The Laboratory's annual budget is

approximately $1 billion [U.S.

dollars]; it's facilities (excluding

utilities and the contents of

buildings) are estimated as being

worth $810 million [U.S. dollars].

1 . 2 Seismic Setting of Laboratory

LLNL is located in the Livermore

valley about 40 miles [64.4 km] east

of San Francisco and Oakland,

California. This site is in the

coastal region of Central

California, one of the most

seismically active regions in the

United States . Figure 1 indicate

the major fault systems which pose a seismic hazard to Laboratory

facilities. To the west (60

kilometers) lies the major San

Andreas • fault zone which extends the

entire state of California. Closer

in, but still to the west, are branches of the San Andreas; the

Hayward and Calaveras fault zones. These latter faults are approximately 25 and 17 kilometers

from the Laboratory, respectively.

And finally, several local active

faults surround the Laboratory

within the Livermore Valley.

The Laboratory has spent some $5M

[U.S. dollars] to characterize the

seismic hazard it faces. -^'^ This

study, started in 1979 and completed

in 1984, included literature

reviews, geological mapping,

drilling bore holes, excavation,

interpretation of aerial

photography, soil dating,

seismological studies and

geophysical surveys. The

Laboratory's work was reviewed by a consulting firm and a three-member

panel of academic experts.

Major conclusive findings indicate:

(1) no evidence of active

earthquake faults underneath the

Laboratory; (2) land slide and soil

liquefaction at the Laboratory are improbable; and (3) the major threat

to the Laboratory is that of strong

ground motion. Further, the

greatest seismic threat to the

Laboratory is from a moderate-size

(M=5.5 to 6.5) earthquake on local

faults in the Livermore Valley.

Major earthquakes on the San

Andreas, Hayward and Calaveras

faults are less of a threat than

local faults due to their distance

from the Laboratory.

Figures 2a and 2b characterizes the

seismic hazard to the Laboratory as the result of the $5 million [U.S.

dollars] study. Figure 2a shows the

expected peak free field ground

acceleration verses the mean number

of events per year. Figure 2b gives

a response spectra normalized to Ig.

These Figures represent the most

current information available, and are used for the design of new Laboratory facilities.

1 . 3 Impact of 1980 L ive rmore Earthquake ^

^

On January 24, 1980, LLNL

experienced an earthquake with a Richter magnitude of 5.8, followed

within three minutes by three

relatively large after shocks with

magnitudes of 5.1, 4.0, and 4.2. On

January 26th, another earthquake

with a Richter magnitude of 5.4

occurred. The epicenter of the

January 24th event was about 13

miles [20.9 km] northwest of the

Laboratory; the epicenter of the

January 26th event was about 4 miles

[6.4 km] northwest of the Lab. The

peak ground accelerations at LLNL

were estimated to be in the range

from 0.2g to 0.3g.

Although no one was seriously

injured, the 5.8 earthquake caused

considerable damage and disruption

at the Laboratory-^. This is the

only earthquake that caused

143
significant damage to DOE facilities

during the past 46 years^ • Building

walls of concrete and masonry were cracked and broken. Extensive

cracking occurred in concrete

structures throughout the

Laboratory. Most of this cracking

was not a threat to the structural

integrity of the buildings or structures. However, Building 311,

a two-story physics concrete-frame

office building suffered

considerable column damage at the

juncture of the second floor and

roof lines, equipment shifted from

mounting pads, and staircases

cracked. Building 113, the

Laboratory's five-story computer

office building, suffered extensive

cracking to its central lateralforce resisting concrete core. Building 111, a seven story

administrative building, suffered

damage to an exterior seven story,

free-standing stair tower.

Extensive cracking was observed in

the central core concrete walls, a second floor soft-story, and at

column to beam connections.

In other Laboratory buildings, some connections between tilt-up concrete

walls and their supporting

structural-steel frames had failed.

A few structural steel welds failed,

several loose or stretched bolts

were noted, and damage to the

connections of structural members

was found throughout the site. A

few structural steel framing members

buckled in some of the buildings.

Many elevators could not be used

because counterweights had been

knocked loose from their mounting

brackets, bending the guide rails.

Surprisingly, the Laboratory had

very little window glass breakage.

However, concrete anchors used to

fasten down various types of

equipment and tanks were twisted or torn and, in some cases, were completely pulled out of the

concrete

.

Trailer support stands moved and

toppled. Many trailers shifted,

causing their outside doors to jam or be blocked, so that they could

not be immediately opened.

The Laboratory spent approximately

$10 million (in 1982 U.S. dollars)

in earthquake restoration, repair,

and upgrades from 1981-1985.

Approximately $7 million [U.S.

dollars] of this was directed toward

seismic tie-down of: office

equipment and library shelving;

mechanical and electrical equipment;

trailers; ceiling and light

fixtures; acoustical tile and lay-in

panel ceilings; mechanical

restraints to cranes;and the

replacement of rigid with flexible

connections to propane tanks.

Approximately $3 million of the $10

million [U.S. dollars] was spent on major building upgrades of Building

311 and Building 113. Another $6.5

million (in 1990 U.S. dollars) is

currently being spent to upgrade

Building 111. All three of these

major upgrades are discussed in

detail later.

2. MAJOR BUILDING UPGRADES

2 . 1 Seismic Criteria for New

Buildings

In general, the seismic design

criteria used for Laboratory

facilities has the following major

safety goals: (1) no loss of life;

(2) no building collapses, and (3) no compromise of containment in

hazardous facilities

.

The seismic criteria used at the

Laboratory reflect four eras: (1)

pre-1971; (2) 1971-1980; (3) 1980-

1989, and (4) present (U.S. DOE

6430.1a6/uCRL-15910.^)

.

144
Pre-197

1

Before 1971 all facilities

were constructed to meet the latest

edition of the Uniform Building Code

(UBC) in effect at the time of

construction

.

1971 to 1980 The 1971 San Fernando

earthquake, (magnitude 6.3), caused

far more damage than expected by

earthquake engineers. As a result,

the Laboratory expended a considerable effort to review its

earthquake design criteria and its

facilities. The result was an upgrade to our hazardous facilities

criteria such that the upgraded

criteria exceeded UBC criteria . No

changes were implemented for

conventional facilities.

1980 to May 1989 All Laboratory

buildings, structures, systems, and

components were categorized into one

of four safety classifications:

High Hazard, Moderate Hazard, Low

Hazard, and No Hazard (Conventional

Building) . Each safety

classification had its own design

basis earthquake criteria. Table 1 summarizes Laboratory seismic

criteria used post-1980 Livermore

through May 1989 for new facilities.

High Hazard structures, systems, and

components (Seismic Category I items) went through a two-step

design process. Seismic Category I structures, systems, and components

are those items whose continued

integrity and/or operability are essential to assure the capability

to shut down and maintain a safe

shutdown condition, and to prevent

or mitigate the consequences of

accidents which could result in

potential off-site exposures. The

first process consisted of

evaluation and design in the elastic

range of response. It was done

using a dynamic response spectrum

analysis (Figure 3) with a corresponding horizontal Design

Basis Earthquake (DBE) peak ground

acceleration (PGA) of 0.50g applied

simultaneously with a vertical PGA

of ± 0.33g. Engineering evaluations

and design were done by using UBC

Strength Allowables. Connection

evaluation and design included an additional load factor of 1.5. That

is, the connection design/evaluation

used forces which were 1 . 5 times

those seismically calculated, . The

second part of the design process

allowed for inelastic response

Seismic Category I items were checked against forces due to a DBE

with a horizontal PGA of 0.80g

applied simultaneously with a vertical PGA of ±0.53g. For this

check, the item was to remain

functional during and after the

earthquake

.

Moderate Hazard structures, systems,

and components (Seismic Category I items) also went through a two-step

design process . The first process

consisted of evaluation and design

in the elastic range of response.

It was done using a dynamic response

spectrum analysis (Figure 3) with a corresponding horizontal peak ground

acceleration (PGA) of 0.25g DBE

applied simultaneously with a vertical PGA of + 0.17g.

Engineering evaluations and design

were done using UBC analysis methods

along with UBC Strength Allowables.

Connection evaluation and design

again accounted for an additional

load factor of 1.5. The second part

of the design process allowed

inelastic response and checked the

integrity of the Moderate Hazard

Seismic Category I items against

forces due to a DBE with a horizontal PGA of 0.50g applied

simultaneously with a vertical PGA

of ± 0.33. For this check, the item was to remain functional during and

after the earthquake

.

Non-Seismic Category I items were designed using LLNL's No Hazard,

Standard Facility Criteria. An item

in a lower Safety Classification

must not cause failure of an item in

a higher Safety Classification.

145
Low and No-Hazard conventional

structures, systems, and components

were designed to the following

guidelines:

1 . One and Two-Story Buildings and

Structures Used current UBC

requirements with a more

conservative seismic base shear

coefficient of 0.25 (static).

Connection evaluation and design,

due to seismic forces, accounted for an additional load factor of 1.5.

To assure that the building's

structural elements could reach

their maximum potential for

ductility

.

Analysis procedures, design

procedures, and material strength

allowables met the requirements of

the latest edition of the UBC and

DOE Facilities General Design

Criteria

.

2 . Buildings and Structures

Greater Than Two Stories As a minimum, buildings and structures

over two stories in height were designed to meet the seismic design

requirements for 1) above.

Buildings and structures over two

stories in height were also

evaluated and designed using a response spectra (Figure 3) with a corresponding horizontal peak ground

acceleration (PGA) of 0 . 5g applied

simultaneously with a vertical PGA

of ±0.33g.

Engineering evaluations and design

was accomplished by the use of

inelastic analysis methods combined

with inelastic stress allowables.

Primary concern was to ensure

prevention of building collapse,

thereby allowing the building

occupants to egress safely following

an earthquake having major intensity

at the site. Major damage to the

structure (perhaps even nonrepairable) following this level of

design motion was permissible.

3 . All Buildings and Strnri-nre ^ with Significant Cross Axes Coup Jing

and/or Torsional Response The

design of buildings and structures

placed within this category met the

requirements of 2) above.

2 . 2 Seismic Criteria for

Building Upgrades

The engineering design approach used

for the seismic upgrade of exist ing

LLNL facilities was somewhat

different than the previously

defined criteria required for "new"

LLNL facilities. For the seismic

upgrade on No-Hazard (Conventional)

and Low-Hazard structures, the basic

premise of the LLNL seismic upgrade

criteria was to ensure "life safety"

in the event of a major earthquake.

That is, the building could endure

major structural and non-structural

damage, but no collapse. The

occupants should be able to exit the

building without incurring loss of

life. In the case of Moderate/High

Hazard structures, systems, and

components the basic premise of the

LLNL seismic upgrade criteria was to

ensure both life safety and

functionality of the Seismic

Category I items. Seismic Category

I items must be able to perform

their required mission. Applicable

codes and standards were consulted

at the time of the upgrade.

The specific seismic upgrade

criteria used on the three major

building upgrades are discussed in

the following sections.

2 . 3 Major Upgrade of Building 311

Building 311 was originally designed

under the requirements of the 1961

Uniform Building Code (UBC) and

constructed during 1964 as a twostory, cast-in-place reinforced

concrete structure with 37,000 ft.^

[3,437 m^ ] of gross floor space (see

Figure 4). It is 95ft. [29m] by 190

146
ft. [57.9m] in plan and consists of

two levels plus a penthouse for

mechanical equipment located on the

roof. The typical bay size is 23

ft. 9 in. [7.239m] by 23 ft. 9 in.

[7.239m]

.

Building 311 's primary lateral force

resisting system was a "non-ductile

moment resisting frame" founded on drilled piers. The first floor is a slab-on-grade, and the second floor

and roof are framed with concrete

beams, girders, and infilled with

concrete pan joists. Building

columns are of reinforced concrete.

There are no column ties within the

Building columns other than in the

immediate vicinity of the ground,

second floor, and roof slabs. The

non-ductile frames were designed

using a uniform lateral force

distribution (1961 UBC) and a base

shear coefficient of 0.067. This

type of framing system has a history

of partial or total collapse in

moderate to strong earthquake

shocks. A significant difference

exists between today's current

seismic building codes for new construction and the 1961 Uniform

Building Code. The 1979 and 1982

editions of the UBC would require

Building 311 to resist a non- uniformly applied static earthquake

force with a base shear coefficient

of 0.094. The 1988 UBC and DOE

Order 6430.1a would require a base

shear coefficient of 0.22 and 0.235,

respectively

.

Building 311 suffered considerable

structural damage during the 1980

Livermore earthquake. This included

vertical column splitting, concrete

cracking, and cracking at the

beam/column joints. Within hours

following the earthquake, LLNL

structural investigation teams were assembled and in the field.

Engineering consulting firms were hired to review Building 311, (as

well as Building 113, and Building

111) . A sign was posted at all

entrances and exits of Building 311

informing both visitors and building

occupants as to the inherent

structural deficiencies (weaknesses)

with respect to the ability of the

building to safely resist earthquake

forces. With the installation of

temporary wood column shoring.

Building 311 was considered safe for

habitation while the Laboratory

proceeded with haste to complete

building repairs. Laboratory

personnel were allowed to re-occupy

the building during the remedial

crack repairs . Concrete cracks were repaired by injecting high-strength

epoxy grout into the cracks, a procedure that can effectively

restore earthquake damaged concrete

structures to nearly their full

design strength.

After the earthquake, the Laboratory

decided to upgrade the seismic

performance of Building 311 with

respect to both damage control and

life safety concerns. The seismic

upgrade criteria required; (1) the use of 1979 UBC design allowables,

(2) a UBC base shear coefficient of

.25, (3) a connection static force

design multiplier (all connections)

of 1.5, and (4) no building collapse

at 0.50g peak ground acceleration

with design response spectra shown

in Figure 3.

The structural upgrade solution was to construct and attach to the

existing building a series of

external two-story-high concrete

shear walls (buttresses) on three of

the four sides of Building 311, two

on the west end, two on the east

end, four on the south side, and

none on the north side, (see Figure

5) . The concrete buttresses are anchored to the ground using 3-ft.

[.914m] diameter by 40-ft. [12.192m]

long drilled concrete piers, four

piers per buttress. The buttresses

are attached to the second-floor

level of the building by rebar

welded to steel-channel drag struts.

These struts are, in turn, attached

to the underside of the second-floor

concrete beams by specially designed

steel bolts (see Figure 6) . The

second-floor connectors use a straight pattern of ten (10) custom147
made bolts spaced approximately 9 inches [22.86 cm] apart on center at

each buttress. These bolts have a length of 16 inches [40.64 cm] with

an overall diameter of 3-3/4 inches

[9.53 cm] for that portion of the

bolt passing through the concrete

beam. The bolts are necked down at

each end to a 1-1/4 inch [3.18 cm]

diameter threaded portion for

bolting to the steel drag struts.

Buttresses are also attached to the

concrete roof slab by the use of

rebar embedded in the buttress and

welded to a steel plate drag strut,

which is, in turn, attached to the

top of the respective concrete roof

beam by 13 specially-made shear

connectors spaced approximately

seven inches on center (see Figure

7) .

A modification to the above design

provided high load capacity drag

strut connections commensurate with

the non-uniform load distribution

across the second floor drag strut

bolts and the roof shear connectors.

This was done since the first drag

strut connecting anchor, adjacent to

a buttress, must resist seismic

forces several times larger than the

last anchor furthest from a buttress. Load distributions to the

anchors were determined by a series

of computer models of typical

buttress/drag strut configurations.

Design variables considered during

this process included: connector

strength at elastic and inelastic

levels, connector stiffness,

concrete bearing, tensile, shear,

and compressive stress levels, and

overall total connector system

stiffness (tension stiffness vs. compression stiffness)

.

The structural upgrade costs of

Building 311 were approximately $1.5

million (in 1982 U.S. dollars)

.

Replacing the building would have

cost approximately $4.6 million (in

1982 U.S. dollars). The

construction time for the structural

upgrade took about 10 months.

2 . 4 Major Upgrade of Ruilding

Building 113 is a five-story

reinforced concrete structure,

approximately 90 ft. [27.43 m]

square in plan dimensions (3 bays by

3 bays), containing about 42,000

ft. 2 [3,902 m2 ] of office space (see

Figure 8) . The building has a partial basement and a 14-ft. [4.27

m] high steel-frame penthouse for

mechanical equipment located on the

roof. Each of the five stories is

13-ft. [3.96 m] high, and the

basement floor is 17-ft. [5.18 m]

below the first floor level. The

basement forms a passageway to the

adjoining computer center building.

The five-story structure is founded

on 24-in. [61 cm] diameter cast-inplace reinforced concrete piles.

The building has reinforced concrete

waffle slabs supported on reinforced

concrete columns, spaced on 30-ft.

[9.14 m] centers each way, and a core of reinforced concrete shear

walls located in the west half of

the structure. The shear walls

enclose two stair wells and elevator

shafts, and are arranged so that the

center of their rigidity falls close

to the geometric center of the

building. The center core of shear

walls were originally designed to

carry about 80% of the lateral

forces acting on the building, with

the remaining 20% being resisted by

the perimeter reinforced concrete

"non-ductile moment resisting

frames .

Building 113 was designed in 1964 to

meet the requirements of the 1961

Uniform Building Code (UBC) . The

original building design provided

for a lateral force base shear

coefficient of 0.057. The 1982

edition of the UBC would have

required Building 113 to resist a statically applied lateral

earthquake force with a base shear

coefficient of 0.14. The 1988 UBC

and DOE Order 6430.1a would require

a base shear coefficient of 0.183

and 0.196 respectively.

148
The January 1980 earthquake caused

no damage to the building's columns

and moment frames, structural damage

was confined to the central-core

concrete shear walls. The walls

developed extensive diagonal tension

cracking and sliding shear cracking

along horizontal construction

joints, which also showed signs of

grinding or spalling. Cracking

occurred throughout the building,

but more extensively in the lower

stories

.

Immediately following the 1980

earthquake, Building 113 was vacated

for approximately 3 days while a structural evaluation of the

building was performed. It was found to be safe to house it's

occupants. As the ground

acceleration of the January 1980

earthquake was estimated to be about

0.25g at the Laboratory site,

therefore. Buildings 113, 111, and

311 were subjected to earthquake

forces at least four times greater

than that provided for in the

original design. High-strength

epoxy grout was injected into the

central-core shear wall cracks,

repairing them in the same manner as was done for Building 311.

By February of 1981, the Laboratory

had begun the engineering effort to

structurally upgrade Building 113

and to increase it's earthquake

resistance with respect to life

safety. The Building 113 seismic

upgrade performance requirements

were as follows:

1. The building shall be capable

of resisting, with no collapse, a

0 . 5g horizontal peak ground

acceleration with LLNL's design

response spectra (Figure 3);

2. Extensive damage at this

earthquake level (0.5g) is

acceptable provided that people can safely exit the building; and

3. The new structural elements

required for preventing the collapse

of the building at 0 . 5g shall meet

the design requirements of the

current UBC . In addition, major

connections shall include an additional safety factor of 1.5.

Many different schemes for upgrading

the building's structure were investigated. In the end, by

considering such factors as cost,

aesthetics, and disruption as being

equal, it was decided that the best

approach was to keep the majority of

the new construction on the outside

of the building. This would allow

the building to remain occupied

during reconstruction. The existing

foundation pile system was considered to adequately meet the

life safety seismic criteria for the

upgrade. Therefore, the upgrade

scheme that was finally decided upon

used steel "K" braced frames

attached, both to the outside face

of the building perimeter, and to

the existing foundation system (see

Figure 9)

.

The "K" brace frame design was based

on the premise that the reinforced

concrete, center core shear walls

will absorb a lot of energy while

degrading during the specified 0 . 5g no collapse earthquake. A damping

value of 10% was used in the dynamic

evaluation for the building's

integrated, composite "K" brace

frame computer model . A

conservative amount of building

period shift was allowed in the

dynamic solution to account for

softening of the structure with a corresponding decrease in intensity

of seismic inertia forces.

Consideration was given for

secondary effects such as P-Delta

contributions. An upper bound of 4 was used for the global ductility of

the steel frame with special

considerations for each structural

element using Newmark-Hall type

procedures. The American Institute

of Steel Construction (AISC) design

procedures for plastic design of

steel members, along with

considerations for member

instability, were used in the design

process. The UBC static design

149
procedures, required for all new structural elements, included a braced frame load factor of 1.25

with no 1/3 increase allowed in

material allowables. The 0 . 5g life

safety design criteria controlled

over the UBC static design criteria

for the design of the structural

elements

.

The "K" brace frames are made of

high-strength steel (Fy=50,000 psi

[3 . 515xl09kg/cm2] ) rectangular

tubular welded sections, with 3/4-

inch [1.91 cm] diameter Nelson stud

anchors welded to their backs (see

Figure 10) . The outside edges of

the building's concrete floor and

roof slabs were chipped back to

expose the reinforcing bars, and the

"K" frames were installed along the

outside perimeter of the building by

attaching the Nelson studs to the

building-slab edges with highstrength concrete grout. The "K"

frames were attached to each other

with welded steel plates, and to the

existing foundation system by

constructing new concrete grade

beams. "K" frames were placed in

all of the first and second story

outside building bays, but only in

the outside middle bays of the

third, fourth, and fifth stories,

thus forming an inverted "T"

configuration at each building face.

The new building frames were designed to safely resist all of the

seismic forces acting on Building

113 as a result of a major

earthquake. Cost of the structural

upgrade was approximately $1.1

million (in 1982 U.S. dollars) . It

is estimated that replacing the

building would have cost about $5.9

million (in 1982 U.S. dollars) . The

construction time needed for doing

the structural portion of the

upgrade was about nine months.

2 . 5 Major Upgrade o f Build ing 111

Building 111 was designed in 1966,

constructed in the late 1960 (s), in

accordance with the provisions of

the 1964 Uniform Building Code

(UBC) . Building 111 is primarily

used as an office building. The

structure was constructed entirely

of reinforced concrete (RC) and

built in a cross-shape plan with

three seven-story wings, one fivestory wing, and a central core area containing elevators, stairs, air

conditioning duct shafts, and

utility conduits (see Figure 11)

.

There are two staircases in addition

to the ones located at the building

core. The first of these is an integral part of the south end of

the north-south wing. The other was originally designed to behave as an independent, free-standing,

reinforced concrete (cantilever

beam) seven-story high stair tower,

immediately to the south of the

east-west wing. The story heights

(floor to floor) are 18 ft. 0 in.

[5.49 m] for the first story and 13

ft. 6 in . [4.11 m] for the upper

stories . The north-south wing is

212 ft. 6 in. [64.77 m] long and 82

ft. 6 in. [25.15 m] wide at the

first floor level. The east-west

wing is 120 ft. [36.58 m] long and

42 ft. [12.80 m] wide. The junction

of the two wings is approximately

1/3 the distance from the north end

of the north-south wing. The width

of the north-south wing reduces to

42 ft. [12.80 m] at the second floor

and remains so at the upper floor

levels . The north end of the

building, beyond the east-west wing,

has only five floors. The building

is founded on cast-in-place

reinforced concrete piles 20 inches

[50.8 cm] in diameter embedded 44

ft. [13.41 m] into the ground, and

contains 102,000 f t

. 2 [9,476 m2] of

gross floor space with a current

population of 323 occupants.

The "vertical load carrying system"

of the building is comprised of 4-

1/2 inch [11.43 cm] thick RC roof

and floor slabs supported using 14

150
in. [35.56 cm] by 24 in. [60.96 cm]

deep RC beams spanning 4 0 ft. [12.19

m] between 14 in. [35.56 cm] by 24

in. [60.96 cm] . RC columns spaced

at 10 ft. [3.05 m] on center along

the outside perimeter of the

building wings. The exterior edges

of the floor and roof slabs are supported by 3 ft . 1 in . [.940 m]

deep by 10 in. [25.4 cm] thick precast concrete spandrel beams topped

by 14 in. [35.56 cm] deep by 17 in.

[43.18 cm] wide RC beams cast in

place with the floor slabs. These

RC beams are connected to the precast spandrels by vertical steel

dowels. The portions of the floor

and roof slabs in the building's

center core area, adjacent to the

interior stairs and elevator shaft, are supported by the RC walls

enclosing the shafts.

The "lateral force resisting system"

as originally designed is comprised

of (1) vertical RC shear walls at

the exterior ends of the south,

east, and west wings; (2) RC walls

enclosing the elevator, stair, and

duct shafts; (3) RC walls enclosing

portions of the five-story wing; and

(4) the transverse beam/column

frames spaced at 10 ft. [3.05 m] on center along each wing (see

"vertical load carrying system")

.

The original building design (1964

UBC) provided for a lateral force,

base shear coefficient of 0.057.

The 1982 edition of the UBC would

have required Building 111 to resist

a statically applied lateral

earthquake force with a base shear

coefficient of 0.14. The 1988 UBC

and DOE Order 6430.1a would require

a base shear coefficient of 0.183

and 0.196, respectively.

As a result of the January 1980

earthquake. Building 111 suffered

diagonal tension cracking to the

central-cores' (RC) shear walls.

Cracking occurred to the RC floor

and roof slabs (diaphragms) in the

vicinity of the buildings' centercore and end-wing, shear walls . The

transverse beam/column frames

suffered damage at the beam/column

joints, and the free-standing seven- story high stair tower was found to

have swayed approximately 3 inches

[7.62 cm] in the north-south

direction, striking the building,

and causing minor damage. The

condition of Building 111 was investigated with respect to

continued habitation, found safe to

house its occupants, and crack

repairs commenced using epoxy grout

injection techniques. The outside

seven-story stair tower was closedoff until a structural upgrade could

be implemented. , ,. ..

By the Spring of 1981, engineering

studies identified the seismic

deficiencies found within Building

111. The list of major structural

deficiencies, based on the original

design of Building 111, include the

following: ' •'• '

•'

1. The foundation of the east wing was not connected to the south wing.

2 . The ends of the west wings have

solid multi-story concrete walls

terminating at a lower level on a concrete frame and thus creating

"soft stories"

.

3. The seven-story cantilevered

stair tower is unsafe and should be

structurally upgraded.

4 . The concrete shear wall at the

end of the east wing has an offset

in it's vertical plane at the second

floor

.

5. The steel reinforcement at the

beam/column joints is not laterally

confined and the anchorage and laps

of bars are inadequate . These

frames will only provide nominal

seismic resistance.

6. The building lacks adequate

shear wall strength capacity.

7. The roof and floor slab

diaphragm cord members lack adequate

reinforcement

.

151
8. The connection of the diaphragm

collectors, cords, and tie elements

(south, east, and west wings) to the

building's center core is

inadequate

.

9. The reinforcement anchorages

for the present spandrel beams is

not adequate

.

The first three structural seismic

deficiencies of the above list have

been corrected. The stair tower is now laterally supported at each

floor level and at the roof by

structural-steel tube connections to

Building 111. High-strength steel

bolts were used to connect the tubes

to the stair tower. Cinch anchors

and grouted steel bolts connect the

tubes to Building Ill's peripheral

beams and columns (see Figure 12) . The second story window openings in

the concrete walls at the west end

of the west wing and the northeast

and northwest corners of the fivestory wing were filled with

reinforced concrete to provide

additional seismic resistance and

the elimination of the "soft

stories". Concrete grade beams have

been added at the first floor

breezeway (now enclosed) at the end

of the east wing adjacent to the

building core.

LLNL is presently in the process of

retrofitting the remaining Building

111 structural seismic deficiencies

at an approximate cost of $6.5

million ( in 1990 U.S. dollars). It

is estimated that replacing the

building would have cost about $32

million [U.S. dollars]. The

construction time required for the

building upgrade is 12 months.

Construction is currently scheduled

to be complete by January, 1991.

The Building 111 seismic upgrade

performance requirements, are as follows

1. The building shall be capable

of resisting, with no collapse, a

0 . 5g horizontal peak ground

acceleration with the acceleration

response spectrum shown in Figure

13. (Note, this response spectrum

was based on the 1988 UBC spectrum

shape .

)

2 . Extensive damage at the

earthquake ground motion level

defined in (1) is acceptable

provided that people can safely exit

the building.

3. All new structural elements

required for preventing the collapse

shall meet the design requirements

of the current UBC. In addition,

brittle-type connections shall

include an additional safety factor

of 1 . 5

.

4. Minimize disruption to the

building occupants during the

construction phase of the

strengthening process.

5. Maximize aesthetics for the

resulting construction while

minimizing total project cost.

A minimum of six solutions were developed and considered in detail

with respect to the seismic upgrade

of Building 111. Considerations for

minimum disruption and project

costs, with the safety of building

occupants being equal for all

proposed solutions, again drove the

upgrade scheme to the outside of the

building, thus, allowing the

building to remain occupied during

the implementation of the upgrade

solution.

Figure 14 in an artist rendering,

illustrating the Building 111

current upgrade. Two large

reinforced concrete (RC) towers will

be located at the exterior of the

building. One tower is located

adjacent to the east wing on the

south side, and the other tower is

located adjacent to the south wing

of the east side. The two towers

rise from a 6 ft. 8 in. [2.03 m]

thick reinforced concrete mat

supported on thirty nine 36 inch

[91.44 cm] diameter drilled RC piers

152
extending 50 ft. [15.24 m] to 66 ft.

[20.12 m] (depending on location)

below the new foundation mat.

Each RC tower is connected to the

adjacent building wing with channel

shaped steel collector members

installed on each side of three

floor beams starting on the third

floor to the roof. Each collector

member extends full width of the

office wings, is securely anchored

into the concrete tower, and is

bolted and bonded with epoxy grout

to the concrete floor or roof beam.

Construction of the collector

members will be accomplished by

removing the precast concrete panels

at four bays in each wing to permit

access. Plywood sheets will be laid

on top of the existing ceilings for

construction contractor personnel to

drill holes in the existing concrete

and install the collector members.

This means of construction will

permit uncleared construction

personnel to perform the work

without penetrating secured building

areas. Epoxy grout is injected

between the steel collector member

and the existing concrete beam along

with having a steel plate anchor at

the exterior end. The horizontal

collector tie beams are constructed

to transmit the collector member

forces into the concrete towers. A

few additional RC walls will be

constructed internal to the building

to help in distributing shear

forces

.

3. CONCLUSIONS

The Lawrence Livermore National

Laboratory experienced considerable

damage as the result of the 1980

Livermore Earthquake. Subsequently,

a major study was undertaken and

completed in 1985 to quantify the

seismic hazard the Laboratory faces

.

The design criteria used for new facilities and upgrades (retrofits)

of existing facilities are consistent with the results of this

study. Based on post-1980

inspection of Laboratory facilities.

by both independent consultants and

Laboratory staff, all buildings

requiring upgrade to assure "life

safety" have been identified. Three

major structural building upgrades

were involved: Building 311,

Building 113, and Building 111.

Upon completion of the Building 111

upgrade, currently under

construction, no further upgrades

are required or planned.

4 . REFERENCES

1. "Final Report, Lawrence

Livermore National Laboratory

Seismic Exposure Comparison Report

and Spectra Report," prepared for

LLNL by Geomatrix Consultants, Inc.

(1985)

.

2. J. F. Scheimer, "Lawrence

Livermore National Laboratory Site

Seismic Safety Program—Summary of

Findings," UCRL-53674 (1985).

3. G. E. Freeland, Lawrence

Livermore National Laboratory

Earthquake Safety Program: UCAR

10129 (1984)

.

4. R. C. Becker, et al, the LLNL

Earthquake Impact Analysis Committee

report on the Livermore, California

earthquakes of January 24 and 2 5,

1980: UCRL-52956, (1980).

5. J. R. Hill, Injury and Property

Damage Experience from National

Hazards 1943 to 1989, Second DOE

National Phenomena Hazards

Mitigation Conference, Knoxsville,

Tennessee (1989)

.

6. U.S. Department of Energy,

"General Design Criteria", DOE Order

6430.1a, April 6, 1989.

7. R. P. Kennedy, et al, "Design

and Evaluation Guidelines for

Department of Energy Facilities

Subjected to Natural Phenomena

Hazards": UCRL-15910 Draft (1988).

8. G.E. Freeland, "Earthquake

Safety Program at Lawrence Livermore

National Laboratory, " DOE Natural

Phenomena Hazard Mitigation

Conference - 1985.

9. G. E. Freeland and Manohar

Sethi, Seismic Upgrade of Building

311, 8th World Conference on Earthquake Engineering, San

Francisco, CA (1984) .

153
L.L.N.L. SEISMIC DESIGN CRITERIA

NEW FACILITIES, 1980 TO MAY 1989

FACILITY USE SAFETY CLASSIFICATION

LOW HAZARD AND NO

HAZARD (CONVENTIONAL)

(1) 1 & 2 STORIES

V= 0.25 W

UBC ALLOWABLES

1.5 CONNECTION

FACTOR

MODERATE HAZARD

(1) CATEGORY I

a) 0.25g RESPONSE

SPECTRA

UBC ALLOWABLES

1.5 CONNECTION

FACTOR

HIGH HAZARD

(1) CATEGORY I

a) 0.5g RESPONSE

UBC ALLOWABLES

1.5 CONNECTION

FACTOR

b) 0.8g RESPONSE

SPECTRA

REMAIN

FUNCTIONAL

b) 0.8g RESPONSE

SPECTRA

REMAIN

FUNCTIONAL

(2) 2 STORIES OR

IRREGULAR

a) SAME AS (1)

b) 0.5g RESPONSE

SPECTRA

NO COLLAPSE

EVALUATION

(2) NON-CATEGORY f

CASE-BY-CASE

(GO TO NO HAZARD

CRITERIA)

(2) NON-CATEGORY I

CASE-BY-CASE

(GO TO NO HAZARD

CRITERIA)

154
Figure 1 Major earthquake fault zones which pose a seismic

hazard to LLNL

.

155
Figure 2a Seismic Hazard Curves for all LLNL site: Peak

Ground Acceleration (g)

.

156
o

Oc_ o

Q Q o

rO OsJ —

(6) N0llVa3n3DDV nvaiD3dS

toM

+Jo

Q) CO

(DCO 2: a 0) o q

CO M

0) O

PC H-i

o <u -H CO P

>-l >i

0) -H

rH 4-> 0) C

U (D U ^

c u

-H ». CO !j>

(1) rH Q

O

O

-Hg -d CO

CO (U 0) -H N -H

0) -H +J W rH -H

CO >H g -H o fd C 4-1

CM

np

-H

fa

157
O Q O

^' rO cJ ~

(B) NOIlVd3n3D0V IVdlDBdS

o

M • +J CTi O CX) 0) CTi

>1

(U (0

COao o

a, -p

CO(D rH

o

-H -d

4-> 0)

fO CO M p

O <H o

«< o

c

-H <U

CO N

0) -rH Q

(0g

^^ O

CO

<uM

P

in -H

158
Figure 4 Building 311, as originally designed and

constructed

.

Figure 5 Building 311, after major seismic upgrade showing

external buttresses.

159
Building 311, Second floor drag-struct anchor

bolts

.
Figure 8 Building 113, as originally designed and

constructed.
Figure 10 Building 113, "K" brace frames being lifted into place ,
Figure 12 Building 111, stair tower upgrade, lateral supports

are attached to each building floor level.

163
o o q ^ ro CM

(B) N0liVd313D3

(0 00 1 U CO

+J CTi O iH

M

CD

O

d, <D ^

-P X5

(1) 0

CO

C CO COP O

ft-d -

CO (D (1) <D CO ^ X! C CO O V -H Cn 6

+J rH P

fd

u

M

O -P

(D +J

O

M 0)

<u -d

a

O 0 CO O N

<; -H 0) c fd o

Cr> 6 O

-H v^ CO O iji 0) c c Q -H

-0

U Cr>rH -H C -H <u 6 -H p -d

CO OiPQ fd -H e

(U fd

g Cr>

C/3

Q )^O P

o\o M-l 13 O -H rH ^ C ^ O ^

00

»H

<DUP

Cn -H

fr4 lVdl03dS

164 /
Figure 14 Building 111, artist rendering of the completed

seismic upgrade (currently under construction)

showing the new external towers (buttresses)

.

165
Attenuation Characteristics of Ground Strains

Induced During Earthquake

by

Ken-lchl TOKIDA*' , Kcllchl TAMURA*^ and Koh AIZAWA*^

SUMMARY

The appropriate eslimation of ground

strains Induced during earthquake Is Indispensable for the seismic design of

buried lifeline facilities such as pipeline

systems

.

The ground strains induced during

earthquake are estimated with use of the dense

instrument array data obtained at the Public

Works Research Institute during past 78

earthquakes, and the empirical formulae of

attenuation of maximum ground strains in terms

of earthquake magnitude and epicentral

distance are proposed.

KEY WORDS : Array Observation System. Ground

Strain, Attenuation Characteristics

1. INTRODUCTION

It is well recognized that dynamic

behavior of lifeline facilities such as tubular piping systems embedded in ground

essentially depends on the dynamic response of

subsurface grounds. The seismic deformation

method, which considers ground strains induced

during earthquakes as seismic effects instead

of inertia forces, was developed and is now in practical use for seismic design of extended

structures embedded in ground [1]. Although

investigations on actual ground strains

induced during earthquake are essential to assess appropriate seismic effects to be

considered in the seismic deformation method,

few studies have been conducted, mostly due to the lack of measured data [2].

This paper presents analysis on the

ground strains with use of the dense

instrument array data obtained at the Public

Works Research Institute (PWRI) during past 78

earthquakes. The empirical formulae of

attenuation of maximum ground strains In terms

of earthquake magnitude and epicentral

distance are proposed by multiple regression

analysis

.

2. ARRAY INSTRUMENTATION AT PWRI SITE [31

Theare are tow local laboratory arrays

called Fleld-A and Field-B in PWRI campus, as

shown in Fig.l. Subsurface geological

condition around the PWRI is almost uniform,

i.e. diluvial sandy and silty deposits with

approximate thickness of 50 m rest on gravel

formations. Shear wave velocity of the upper

and lower diluvial deposits is approximately

250 m/s and 400 m/s, respectively.

Fig. 2 shows the instrumentation at Field-A and Field-B. 13 three-components

accelerometers are Installed at Field-A, that

Is 3 on the ground surface, 5 at the depth of

2 m and 5 at the depth of about 50 m, along a cross shaped configuration with each length of

100 m. 6 three-components accelerometers are Installed at Field-B, that is, 1 at the depth

of 2 m, 4 at the depth of 50 m and 1 at the

depth 96 m, along a L shaped configuration

with the length of 100 m and 50 m. The

direction of the cross configuration as well

as the direction of the sensors Is oriented

along north-south and east-west directions.

Signals from 19 accelerometers at both

Fleld-A and Field-B are simultaneously

transmitted by cable to the central processing

room, where the signals are digitized with a time Interval of 1/100 second by 12 bits AD converters. The observation was started in July, 1979.

3. CALCUIJVTION METHOD OF GROUND STRAINS

A tetrahedron consisting of 4 obsevation

pointsd, j, m and p) as shown in Fig. 3 is considered to calculate ground strains.

According to a standard three dimensional

finite element analysis procedure [4], the

ground displacement u{t), v(t) and w(t) in x(East-West) , y (North-South ) and z{Up-Down)

directions, respectively, in the tetrahedron

are assumed to be linear as

u(t)= rt 1+ a 2 x+ a 3 y* a A z v{t)= a a o X* a 7 y+ a B 7. (1)

w(t)=ao+a loX+a iiy+a isz

«I Head, Ground Vibration Division, Earthquake

Disaster Prevention Department, Public Works

Research Institute, Ministry of Construction.

•2 Senior Research Engineer, ditto.

•3 Research Engineer, ditto.

166
in which a ,(1=1 ~ 12) represent constants. Determining a i by prescribing coordinates of

the four observation points 1, J, m and p, Eq.(l) can be written in the form

u(t)

1 v(t)

6V

.w(t)

H

(2)

Ui(t) Uj(t) Ur«(t) Up(t)

V,(t) Vj(t) V^(t) Vp(t)

W,(t) Wj(t) Wm(t) Wp(t)

ai+biX+Ciy+di

z

aj+bjx+cjy+dj z am+bmX+Cmy+

ap+bpX+Cpy+dp z

where

Xk. yk. Zk (k=i, j, m, p): coordinates of

k-th observation point

UH(t), VK(t). WK(t) (k=l, J, m. p):

ground displacement at k-th observation point

3; =

xj yj zj Xm ym Zm Xp yp Zp

C ;= - Xj 1 Xm 1 Xp 1

ZjZm

Zp

br

1 yj zj

1 ym Zm

1 yp Zp

xj yj 1 Xm ym 1 Xp yp 1

(3)

6V

[B, . Bj. B^. Bp]{ 5 }

where

[Bh

{ <5 } =

{ 6 k} =

bn 0 0

0 Ck 0

0 0 dK

Ck bn 0

0 dk Ck dK 0 bK

5 , 5 i 5 m

8 p

Uk

Vk

Wk

(6)

(k=i. j. m. p) (7)

(k=i. j. m. p)

(8)

(9)

It should be noted here that tlie ground

strains { £ } estimated by Eq.(5) represent

the average ground strains in a tetrahedron.

4. CALCUIJVTION OF GROUND STRAINS DURING

EARTHQUAKE

Other constants ak. bk, ck, and dk (k=j,

ffl. p) can be obtained bychanging the subscript

in the order of j, m and p.

6V=

1 Xi yi Zi

1 Xj yj Zj

1 Xm ym Zm

1 Xp yp Zp

(4)

Representing ground strains as,

{ £ }

=

£ X

£ V

£ z

7 xy

7 vz

7 z X

3 u/ 3 X

9 v/ 3 y

3 w/ 3 z

3 u/ 3 y+ 3 v/ 3 X 3v/3z+3w/3y

3 w/ 3 x+ 3 u/ 3 z

(5)

substitution of Eq.(2) into Eq.{5) gives

{ £ } = -^lB]{ 5 }

The array data have been obtained at PWRI site during past 100 earthquakes between 1979

and 1989 as shown in Table 1. Among those

data, the data obtained during 78 earthquakes

with the Japan Meteorological Agency (JMA)

magnitude of 4.0 or greater are used for this

analysis

.

Fig. 4 shows acceleration time histories

recorded at A2C0(GL-2m) and A46C0(GL=-4Gm)

during the earthquake of February 27, 1983

(EQ-28), with a JMA magnitude of 6.0 and an epicentral distance to the site of 22kiii. The

amplification ratio of maximum acceleration

A2C0/A46C0 is 2.0 and 2.6 for N-S component

and E-W component, respectively.

Fig. 5 shows displacement time histories

of the ground for EQ-28, which are calculated

by the double integration of acceleration

records. Acceleration records are integrated

frequency domain with the lower and higher

cut-off frequency of 0.2 Hz and 20 Hz. The

amplification ratio of maximum displacement

A2C0/A46C0 is 1.4 and 2.0 for N-S and E-W component, respectively.

8 tetrahedrons are formulated to calculate ground strains at Field-A as shown

in Fig. 6(a) and 6(b), that is 4 upper-side

tetrahedrons for calculating upper level

ground strains and 4 lower-side ones for lower

level ground strains. 2 tetrahedrons are

167
considered to calculate lower level ground

strains at Fleld-B as shown in Fig. 6(c).

Fig. 7(a) and (b) show time histories of

the upper and lower level ground strains at Field-A calculated for EQ-28, respectively. As seen from Fig. 7, the maximum value of upper

level normal ground strains (ex, e y) Is (100~ 200)x 10"*^ and is larger than that of

the lower level normal ground strains, which

is about 50 x 10~®. The maximum value of upper

level shear ground strain ( r w) is about 100

X 10"*^ and is lager than that of lower level

shear ground strain, which is about 50x10"°.

5. ATTENUATION CHARACTERISTICS OF GROUND

STRAINS

characteristics of maximum ground strains.

A in Eq.(lO) is assumed to be 30 l<m. The

empirical formulae of maximum ground strains

are written as

^

] = a X lO'^'^x ( A +30)^ (11)

where

£ : Maximum normal strain

1 : Maximum shear strain

Table 2 shows the coefficients a, b and c obtained by multiple regression analysis. From Table 2, the attenuation formulae of the

maximum ground strains are obtained as

It is very important to estimate seismic

effect properly In the practical design of

structures. Up to the present, not a few

attenuation equations of peak ground motions

(acceleration, velocity and displacment) have

been proposed and they are applied to the

seismic design of structures.

In the past analyses on attenuation

characteristics of maximum ground motions and

absolute acceleration response spectra, based

on accelerograph records, the following

empirical formula is often used as a practical

formula [5]

.

X = a X 10^"^ X { /I * Aa)^ (10)

where

X : Maximum acceleration, velocity and

displacement / absolute

acceleration response spectral

amplitude

M : Magnitude of earthquake

zi : Epicentral distance [km]

zi Constant to adjust X for small

epicental distance

a, b, c : Coefficients

The attenuation characteristics of ground

strains on the horizontal plane e x, e v and

7 xy, which are Important to be considered in

the seismic design of underground stractures,

are discussed in this analysis. Since the peak

values of the ground strains are different

among the tetrahedrons, the average of the

peak gro'jnd strains over the 4 (Field-A) or 2 (Field-B) tetrahedrons is defined as the

maximum ground strains. Furthermore, larger

value of £ X and e y is defined as the maximum

normal strain £ . The same expression with Eq.(lO) is

assumed to represent the attenuation

Upper Level ground strains at Field-A:

£ =1.237 X 10°- ^°3^^x ( zi +30)-°- X 10-°

r =0.894 X 10°- '^^^'^ X ( A +30)""°- "'''^ X IQ-"^

Lower Level ground strains at Field-A:

£ =1.285 X 10°- ^"^"^ X ( zi +30)-°- X 10"®

r =1.549 X 10°- 2®3^x ( zl +30)-°- ^'°x lO-""

Lower Level ground strains at Fleld-B:

£ =1.506 X 10°- ='=s'^x ( A +30)"°- ^*^^x 10"®

7 =4.860 X 10°- ( ^ +30)-°- =°®x 10"^

(12)

Fig. 8 shows the attenuation of maximum

ground strains e and 7 calculated by Eq.(5)

for each event, comparing with predicted value

by Eq.(12). The followings are pointed out from Fig. 8. 1) The maximum normal upper level ground

strains and lower level ground strains at Field-A are distributed in the range from

5 X 10"^° to 200 X 10"° and from 4 x 10"° to 60 X 10~°, respectively. The maximum shear

strains are distributed In the range from

5 X 10"° to 350 X 10"° and from 4 x 10"° to 100 X 10"° for the upper level ground

strains and lower level ground strains,

respectively. The maximum normal and shear

lower level ground strains at Field-B are distributed in the range from 3 x lO"® to 70 X 10"° and from 6 x 10"° to 100 x 10"°,

respectively.

The maximum shear ground strain Is larger

than the maximum normal ground strain

calculated from the array data and the

upper level ground strain Is larger than

the lower level ground strain.

2) According to the empirical attenuation

equations of ground strain at Field-A, the

coefficient b, which represents the effect

168
of earthquake magnitude on the maximum

ground strains, of the upper level ground

strains Is larger than that of the lower

level ground strains. The coefficient c, which represents the effect of epicentral

distance on the maximum ground strains, of

the upper level ground strains is smaller

than that of the lower level ground stains.

Those facts indicate that the upper level

ground strains are more sensitive to earthquake magnitude and attenuation rate

of the upper level ground strains with

epicentral distance is larger, as compared

with the lower level ground strains.

3) Comparing the empirical formulae of

attenuation of the lower level ground

strains for Fleld-A with that for Field-B.

the coefficient b is almost same, and the

coefficient c for Field-A is a little lager

than that for Field-B.

4) Compared with the coefficient c of the

attenuation equations of maximum ground

accelerations based on SMAC accelerograms,

which Is -1.2 to -1.3 [5], the coefficient

c for maximum ground strains Is larger.

This means that attenuation of maximum

ground strains with epicentral distance is smaller than that of maximum ground

accelerations

.

6. CONCLUSION

The ground strains induced during

earthquakes were evaluated by a finite element

method, with use of the dense Instrument array

data obtained at the Public Works Research

Institute. The empirical formulae of

attenuation equation of maximum ground strains

(Eq.(12)) were presented by multiple

regression analysis based on the array data of

78 earthquakes.

The result of this study might be

regarded as basic information for assessing

the ground strains during earthquakes,

however, it should be noted that those results were derived from the data recorded by

relatively small ground motions. The

accumulation of strong motion records and

further investigations should be encouraged.

ACKNOWLEDGEMENTS

This research was motivated based on the

array data which has been collected for years

under the direction by Dr. T. Iwasaki, Dr. Y. Sasaki, T. Arakawa and Dr. K. Kawashima. The autors would like to express their sincere

gratitude to them.

REFERENCES

[llPublic Works Research Institute, "A Proposal for Earthquake Resistant Design

Method", Technical Memorandum of PWRI,

No. 1185, March, 1977(in Japanese)

[2]Arakawa T. , Kawashima K. and Tamura K.,

"Finite Ground Strains Induced during

Earthquake for Application to Seismic

Design of Underground Structures", Fifth

International Conference on Numerical

Methods in Geomechanlcs , April, 1985

[3lOhkubo T., Iwasaki T. and Kawashima K., "Dense Instrument Array Program of the

Public Works Research Institute and

Preliminary Analysis of the Records", Proc.

of 13th Joint Meeting U.S. -Japan on Wind

and Seismic Effects. U.J.N.R.. May, 1981

[4lZienkiewlcz, 0. C. "The Finite Element

Method in Engineering Science",

McGraw-Hill, 1971

(5]Kawashima K., Aizawa K. and Takahashi K.,

"Attenuation of Peak Ground Motion and

Absolute Acceleration Response Spectra",

Proc. of Eighth World Conference on Earthquake Engineering, July, 1984

[6]Tokida K. , Tamura K. and Aizawa K.,

"Analysis on Strong Ground Motions during

Earthquakes Observed by Dense Array at PWRI - Attenuation Characteristics of Ground

Strains Induced During Earthquake - ", Technical Memorandum of PWRI, No. 2798,

August , 1989 ( in Japanese)

[7]Sasaki Y., Tamura K. and Aizawa K.,

"Analysis on the Finite Ground Strains

Induced During Earthquakes", Proc. of Third

U.S. -Japan Workshop on Earthquake Disaster

Prevention for Lifeline System, U.J.N.R..

May, 1989

169
Tabic 1 Earthquakes Observed at PWRI

MO. EPlCEMIEn JHfl HR6. EPICEM. OniE TIME nEGlO'l LOM. LBT D 1 5 1 KM 1

.

nc p ru 1 kit t Utr 1 n 1 Ml 1

1 DH 1 DHriH(\ 1 r nti 1 39' 50' 0' 36' 9' 0" 50 4 1 22

2 Id's. II. l3'0'<:j Ur r IdHMMIsI r nLr 14 1 0 ' 0" 36 4 1' 0" 90 7 . 1 0 4

1 Q7^ I ? 1 ^ 1 ?i ?7i ?

1

Ltl'IrlHl, LMIDH ll.tr 1 4 0' 2

2 ' 0" 35' 31' 0" 60 ^ 2 72

1 Q7Q 1^ IT nRt 0^1 l?'3.i^.l' UO'^JI^S C005T OF IBMRfiKl FPiEF 1 4 0' 3 50" 36' 28' 0" 60 4 . 3 60

5 IIORIHERM HRMTO 1 39' 5

1

' 0" 36' 6' 0' 50 4 . 2 20

6 1980. 04. 22 14i 34i 16 5 OFF CHUBU 1 37' 55' 0' 32' 9' 0" 400 6 .

6

484

7

8

1960.05.07 03i07i36 IIORIHERM KRtllO 1 39' 57' 0" 36" 4' 0" 50 4 .

0

1 3 1980.05. 12 01 '01 '03 MORIHERM KFiUiO 14 0' 4' 0" 36' 1 2' 0" 60 4 . 4 9

9 1980.06. 16 I6'25i08 TOKiO BBi RE&IOM 14 0' 1 ' 0" 35' 38' 0" 80 4 . 6 55

1

0

1980.06.29 16i20>05 IHU FEU RE&IOM 139' 14' 0" 34' 55' 0" 1 0 6. 7 154

1 1 1980. 03. 24 04i 10' 23 50U1MERII KBNIO 1 39' 46' 0' 35' 58' 0' 80 5. 4 30

1

2

1980.09.25 02'54i23 SE COn'jl OF KRIIIO 1 4 0" 1 3" 0" 35' 31 ' 0" 80 6. 1 69

1 3 1980.09.25 02'54i23 5E COflSI OF KftlllO 140' 13' 0' 35' 31' 0" 60 5. 1 69

1 1 1961.01.26 12'47'44 IIORIHERM KPIIIO 139' 51' 0" 36' 10' O' 50 5.0 2 1 15 1981.03. 12 04; 16:41 SOUIHERII KRIIIO 140' 16' 0" 35' 54- 0" 60 4 . 7 30

16 1981.09.02 18:24:44 E OFF KRMIO 141' B' 0- 35' 46' 0- 40 5. 8 102

1 7 1981.09.03 21:52: 12 MORIHERM KRIIIO 139' 48' 0' 35' 10' O" 60 3.5 25

16 1981.09. 14 05'24:08 IIORIHERM KRIIIO 140' 7' 0" 35' 12' O" 70 4.6 10

19 1981.10.14 06:35:41 IIORIHERM KfilMO 139' 56' 0" 36' 3' 0' 50 3.5 15 20

21

1981.11.30 09:27:00 IIORIHERM KRIIIO 139' 54' 0- 36' 4- 0" 50 4. 4 17 1982.03.07 08: 14:36 E C0fi5I OF KRIIIO 1 4 0' 39' 0- 14 1' 57' 0- 36' 28' 0" 60 5.5 64

22 1982.07.23 23:23'5I E OFF KRIIIO 36' 1 r O' 30 7.0 169

23 1982. 10.25 00i50:01 E C0B5I OF KRIIIO 1 4 0' 3 1

' 0" 35' 54' 0" 40 4.2 47

24 1982. 12.08 05'45'47 IIORIHERM KRIIIO 1 39' 53' 0" 35' 5' 0' 50 4.2 1 7 _25_

26

1982. 12. 16 08i55'53 IIORIHERM KRIIIO 139' 55' 0- 35' 2' 0' 50 3.9 1 7 1982.12.29 19i31i03 IIORIHERM KRIIIO 1 39' 59' 0' 36' 12' 0' 50 4 . 0 1 2 27 1983.01.06 03' I8'59 MQRIHERII KRIIIO 1 39' 54' 6" 35' 5' 1 2" 52 4 . 9 1 6 28 1983.02.27 21 1 14i21 50'JIHERM KRIIIO 14 0' 9

'

1 6" 35' 55' 12" 72 5 . 0 22

29 1983.04.24 20i21'55 MORIHERII KRIIIO 1 39' 56' 24' 36' 1 ' 54" 55 4 . 4 1 5 30 iQfl^ fi'^ r\A irttCi'^tAi 1305. U>.U* lU'Uj'fl' MORIHERM KRIIIO 139' 5 3' 30" 35* 4 ' 35" 55 ^ 3 1 7

3 1 l3u3.U<.U<; Uf'Uj'l*: t LUHdI Ur 3 lUMUKU 14 1 1

1

' 30" 35 54 ' 1 6" 54 ^ n J . O

1 3 7 1 J £ 32 i^ifl^ l30J.U3.Uin? i?*n'^i47 I^.UJ'*'' t Ur r f\HI 1 1 U 14 1 1

' 6 35 4 0' 0 4 9 5 . 2 1 04

3 3

t1^03. Qn^ in lit ^Ai nft lU. Ij II'JO'UD iinmurmi k oil in IjUnlnt'iM nHlllu 1 39 5

3

' 2

4' 35* 5' 1 6" 56 1 7 34 llUntntl'll '^HIItU 140 1

' 24" 36 1 2' 1 2" 60 c 1 J . 1 1 0

I yo J. lu.^o i^ioi'ifl IIORIHERM KRIIIO 1 39' 59' 48" 36' 13' 0" 56 ^J n. 0 1 3 36 lon^ 17 11 n.iiC5.7p MORIHERII KRIIIO 1 40' 7' 36" 36' 15' 6" 56 4 . 3

1 c

1 7

^ 37 IIORIHERM KRIIIO 14 0' 1

' 0" 36* 13' 0" 52 3. 6 1 2 36 1 983. 1 2. 3'J 1 1 1 o'J: 4U E C0R5I OF KRIIIO 1 4 0' 45' 0" 35* 40' 30" 52 5 .

3

79

39 lyofl.Ul.Ul ID1U3141 5 OFF K 1 III; 1 1 36' 50" 30" 33' 37' 1 2" 388 7 . 3 405

40 1984.01.02 18'22i27 MORIHERII KRIIIO 1 39' 54 ' 0" 36' 5' 1 8" 52 4 . 1 1 6

< 4

1

1964.01.09 19'23'55 IIORIHERM KRIIIO 14 0' 1

' 48" 36' 12' IS" 59 3 .

6

1 1 42 1984.01 . 17 20: 1 3i 41 E OFF KRIIIO 14 1' 1 4' 46" 36' 26' 54" 43 5. 6 1 1 1 4 3 1984.01 . 18 00' 31 ' 55 E OFF KRIIIO 1 4 r 1 6 ' 0" 36* 26' 42" 43 5. 9 1 1 3

4 4 1964.02.21 20'51'52 IIORIHERM KRIIIO 14 0' 7

• 4 2" 35* 6' 0" 79 5.0 6 45 1984.03.06 1

1

1 17i 20 IIERR lORlSHlHfl 139' 12' 16" 29* 20' 24" 452 7. 9 758

< 46 1984.03.25 07125' 12 IIORIHERM KRIIIO 140' 2'54" 36* 5' 0" 55 3.5 6

4 7 1964.05.23 07i38i 12 MORIHERM KRIIIO 139' 51' 12" 36' 8'54- 61 4 . 6 20

46 1984.05. 30 09' 491 37 IIORIHERM KRIIIO 1 39' 56' 36" 35* r54" 54 4 . 2 1 6 49 1964.05. 30 02i45:25 MORIHERII KRII10 139' 51' 42" 36' B'42" 61 5.0 1 9 50 1984.05. 30 07i09'57 IIORIHERM KRIIIO 1 39' 49' 30" 35' 9' 30" 60 4. 3 23

51 1984.07.25 13'52il9 IIORIHERM KRIIIO 14 0' 316" 35* 1 r 0" 51 4.8 7 52 1984. 1 1. 1 1 21' 33' 18 MORIHERII KRIIIO 140' 4'36" 36* 11' 36" 55 4.0 8 X 53 1984. 12.01 22i 15:56 IIORIHERM KRIIIO 1 39' 52' 54- 36* 4' 12" 51 3.7 19 X 54 1985.02.03 ie'20i21 SW IBRRRKl PREF 1 39' 50' 24" 36* 10' 24- 56 3.8 22

55 1985.03.20 14:53:50 SH IBflRHKI PREF 139' 53' 18" 36* 4'24- 54 4 . 6 18 56 1985.05.22 15i I6'43 5H IBRRRKl PREF 1 39' 56' 35" 35* 1 ' 42" 56 4 . 3 1

6

57 1 :10 3. U / .II 23: do: 1 y 50UIHERM IBRRRKl PREF 14 0* 17' 30" 35* 56' 42" 45 4 , 2 28 X 58 1 yOD. UtJ.IU 13'13'4/ 1 II n^i ir^i II

1

59 IPR'^ nR 13 IO*^R*n7 HI 1 RCIROk 1 PnFF 1 DllnHlS 1 rnLr 139 54 30 35 9 6 ——

56 4 . 0

1 c

1 J 60 1985.09.12 09'33i07 innpDKl 1 unnHr\ 1 pnFF r iicr 50 4 . 1 1

6

6

1

I9RS 10 04 21 2S' ^2 ^nilTHFRtl jUUIMcnH IDHIinivl 1 RQnni' 1 prff rncr 14 0 9 30 JiJ£ D

7 fl 6 . 1 29

62 1I3UJ.II.^C 9fl'^ It 2? 1I 3lJ ' 1

I 7''Ul 1 ni 1 RnnQK 1 pnFF 1 J y JO jD JD 1 5 4 c 7 J£ 4 9 1 4 63 19R6 n? 12 11 t^Q' 37 r ncc 1 Dnnok 1 ddcc t Ur r 1 OHMHK 1 rntr 14 1' 5* 0" 36* 25' 0" 4 4 1 QRyo

6 4

1 QRfk 1^ nfit^3i73 SW IBRRRKl PREF 139' 52' 18" 36* 9' 6" 57 4 .

0

1 9

D J

1 QQC nc 7il 1 1 • C 3 1 1 n 1 yob. UO. 24 1 1 1 i>3l lU 5E OFF B050 PEUIIISULR 140' 4 3" 12" 34* 49' 24" 73 6.5 156 66 1986.09.20 12i04i56 IIORIHERM IBRRRKl PREF 140' 39' 36" 36* 28' 24" 56 5.0 65

67 1966. 11.15 15i05i 44 E OFF IBRRRKl PREF 140' 55' 42" 36* 23' 42" 43 5.0 82

66 1986. 1 1 .29 07' 29' 35 E OFF IBRRRKl PREF 14 1' 10' 48" 36" 23' 30" 42 5.8 104 69 1987.02.06 22' 16' 15 E OFF FUKU5H1HH PREF 141' 53' 46" 36' 57' 42" 35 6.7 188 X 70 1987.02. 12 04:52:04 UllKMOH

7

1

1987.02.22 05' 39:02 SH IBRRRKl PREF 1 39' 46' 54- 35' r 30" 90 4. 4 29 X 72 1987.02.23 O8i05i03 5W IBRRRKl PREF 14 0' 6 ' 6" 35' 12' 30' 52 3.9 1 1 1987.03. 10 09'37'45 IIORIHERM IBRRRKl PREF 1 40" 43' 24" 36' 23' 54" 65 4 . 8 7 1 X 74 1987.03.25 18i32iOO SH IBRRRKl PREF 14 0' 3 ' 1 2" 36' 1 r 54- 59 3.6 9 75 1987.04.07 09i 40:43 E OFF FIIKUSHIHB PREF 141' 52' 0" 37* 18' 0- 44 6.6 207

76 1987.04.10 19'59i40 5W IBRRRKl FREE 1 39' 52' 30" 36* 7' 12" 51 4.9 16 77 1967.04.23 05i 13:23 E OFF FUKIJSHIMR PREF 14 1' 37- 36" 37' S'lB" 47 6.5 175 X 76 1987.05.26 00131: 17 SH IBRRRKl PREF 139' 50' 36" 36' 10' 0" 59 3.7 21 79 1987.05.11 16'00'32 5W IBRRRKl FREE 140' 2'54" 35' 10' 48" 55 3.9 8 60 1987.06.30 18:17:08 | 5W IBfllWHI PREF 14 0' 5

• 1 8" 36" 11' 0" 57 4.9 7 MOIE 1 LOCBIIOU OF PWRI IS RS FOLLOWS. 140* 4.5'E. 36* 7.4'N

170
Tabic 1 FnrlhqunkcB Obncvod nt rWR[ (continued)

EriCEIIItn jHn EFICEII.

NO. DfllE HUE nE&ioii LON. LRl

.

DEPIHIKHl HPG. 0 1 S 1 KM 1 81 1967. 07. 05 I6'58i06 cil IDnnnir 1 rpcr

1 4 0' 4 ' 4 2' 35' 1 2 ' 1 2" 56 3. 6 9 62 1 ooT m t ^ 1 a I ? 1 •

1 yo /.U ' . I < 1 3' 31

jU CLi iDnnni/l rnir 1 0'""'^ 1 r (Vti 1 4 0 3 ' 36" 36' 6 ' 4 2" 57 4 .

6

3 63 I 39 ' . U ' . * ' Id' Uo' 4

/

1 RflnnK 1 rnPF ^c* in' I 30 1 U 1 £ jD 4 . 1 rJ

6 4 cu ibQno^' 1 mc r J" ItJHIlHf.l rMtr 139 55 30 36 4

' 1 8" 5 1 3 . (

1 c

1 3 65 iyo/.U3.<4 131331^1 F nF F InnnnKl FnFF t Ur r lD'tnMi\| rl\tr 30 3 / 3 o

i 1 11 £ 3 66 iJO 1 • } I . £3 11' 4U" UtJ 14 0' 3' 1 8' 36' 12' 0" 54 3 . 3 Q3 87 1987, 12. 10 lo' 32' 45 5H 1 BnRni; 1 pref 14 0' 0*42" 36' 12' 4 2" 59 4 .

0

1

2

68 1997. 12. 17 1 1 109' 1

7

HUjw'JRl COOjI B050 FEIt 14 0' 29' 4 6' 35' 22' 18" 58 6. 7 92

89 1966.02.07 16'59i2S 5H iBnnnKi rnEE 14 0' 0' 54" 36' 6' 48" 53 3. 5 5 90 1988.03. 16 051 34i29 lOKio rriEF 139' 38' 4 8" 35' 39' 42' 96 6.0 64

31 1988.07. 15 09i5l 5H lonnnKi pp.ef 14 0" 6' 0" 36' 14' 0" 62 4.5 13 92 1988.09. 16 03' 19 SH lenmKi rr.EF 140' 5; 0" 36' 12" 0' 59 4 .

4

9 93 1988. 10.31^ on 25 5M iBnnriKi rnEF 14 0' "7

' 0" 35' 1 r 0" 56 3.8 8 94 1988. 12. 28 18' 02 SU IBRRriKl FBEF 139' 55' 0"" 35' 4' 0" 53 4 ,

3

1 4 95 1989.02.19 21'27 SH isnnmu fref 139' 55' 0' 36' 0' 0" 54 5.6 20

96 1989;03.06 23'39 IIOnillEBIJ CMIBn FREE 1 40" 44' 0" 35' 42' 0" 61 5.9 76

97 1983.03.11 16M2 SQuiMEFji iBnnriKi pref MO' 35' 0" 35' 55' 0" 42 4.9 51 96 1989.03. 17 20'52 511 IBHRnKI PREF i39'56' 0" 36' 6' 0' 58 3.9 1 3 99 1989.04. 12 04:41 5H lenRflKI PP.EF 139' 5r 0"^ 36' 5' 0' 58 4.2 16 100 1989.04.26 02' 18 IIORIMEnil CHlBn PP.EF \'iO' 29' 0" 35' 54' 0" 65 5. 3 44

(lOiE ' LOcniioiJ or pmii 15 ns fniL0n5. ucr i.S E. 36' 7.4 -h X : unused In tlils analysis

Table 2 Cocfriclcnts oT Attenuation Equation

Field Strain

Coefficient Correlation

Coefficient

Standard

Error Numbera b c of Data

A

Upper Level

Ground Strain

e 1.237 0.493 -0.741 0.829 0.206 65

7 0.894 0.548 -0.774 0.856 0.210 65 Lower Level

Ground Strain

e 1.285 0.309 -0.370 0.708 0.213 74

7 1.549 0.293 -0.319 0.699 0.212 76

B Lower Level

Ground Strain

t 1.506 0.358 -0.569 0.736 0.197 66

7 4.860 0.312 -0.596 0.690 0.190 53

Attenuation Equation : ^ 1 =a x 10'^'^ x ( + 30 1"" x 10""

Figure 1 l/)cal I^iboratory Array at PWRI

I 171
425? '

50ni

4 46 CO

t * I

I ILOWION IN,S,E,W )

IDE PTH FROM SURF4CE [m) FIELDA

(a) Field-A

50m

(b) Ficld-B

Figure 2 Array Configuration at PWRI

Figure 3 Tetrahedron for Calculation of Ground Strains

172
100

'^"L" 50 r 11 I. , N-S COMPONENT

G -50

100 L

I I I I I I I I I I I I 1 I I I I I I 1 I I I I I 1 I I 1 I 1 1 I 1 I I I I 1 I I I (5fQ)

0 10 20 30 40 50 60 70 80

100

^^"L'

, 50 E-W COMPONENT

0

-50

100

1 I 1 I 1 1 i_J I 1 I I 1 I I 1 I 1 1 1 1—1 I I I I I I I 1 I1 I 1 1 1 \ i i_J ( 5e; c

)

0 10 20 30 40 50 60 70 80

50 U-D COMPONENT

-50

100

1_J L_J 1

I I I

I I I 1 1 1 t I I 1 1 1 I ^ III I I I I I I I I I I I ] ( 2^ J-

)

0 10 20 30 40 50 60 70 80

(a) A2C0

lOOr^^^L)

50 N-S COMPONENT

0

-50

100

I I I I I I I I I I I I I I I I I I I I I I I 1 I 1 1 I I I 1 1 1 I i_j I I 1 I 1 1 (5f;(;) 0 10 20 30 40 50 60 70 80

100

50 P E-W COMPONENT

-50

100

I I I I I I I I I I I I I I I I I I I I I I I I I I I I I 1 I I I I I I I I I I I I (5e;[;) 0 10 20 30 40 50 50 70 80

100 '^"L)

50 U-D COMPONENT

-50

100

I 1 I I 1 I I I I I I I I I I I I I I I I 1 I I I I II I I I I I I I L-J 1 1

I I I (5EC)

0 10 20 30 40 50 60 70 80

(b) A46C0

Figure 4 Tl«e History of the Obscved Acceleration

(A2C0 and A46C0, EQ-28)

173
l.Oh n N-S COMPONENT

0.0

-1.0

2. 0 L

I I I I I I I I I I I L__l I I I 1 I I I I I I

I I I I I I I I I I I I I I I I I I I ( ) 0 10 20 30 40 50 60 70 80

2.0'^"'

1 .

0

0.0

1.0 - 2.0L

E-W COMPONENT

'^vV^v « .^ '> . ^Ar-.>^.—-~ — —

I I I I I I I I I I t_J I I I I I I I I I

I I I I I I I I I I I I I I I I I I i_j I (5e;[;) 0 10 20 30 40 50 60 70 80

2.0''^"'

l.ol- U-D COMPONENT

0.0

1.0

2.0

I I I I I

I I I I I I I I I I I I I I I I I I I I I I I I I I I I 1 1 I I (

)

0 10 20 30 40 50 60 70 80

(a) A2C0

ICMI 1.0

0.0

1.0

N-S COMPONENT —s{\^'^^^^w^AJ^-—-—

1 I ] I I I I I I I I II I I I I I I I I I 1 I I I I I I I I—I—1-J—l—J—I ( c;^Q ) 0 10 20 30 40 50 50 70 80

(CM)

1 . 0 h 0.0

-1.0

-2.0

E-W COMPONENT

I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I 1 I I I I I I I I ' (SEC)

0 10 20 30 40 50 60 70 80

2 0'^^'

l.of U-D COMPONENT

0 . 0 — .. ...^..^„ —vv--- —— 1.0-

2.0L

I I I I I I I I I I I I I 1 I I I I I—I I I I I I—i_i_J—L_i—I (5EC)

0 10 20 30 40 50 60 70 80

(b) A46C0

Figure 5 TI«e History of the Calculated DisplaceBcnt

(A2C0 and A46C0, EQ-28)

174
(a) Upper Level Ground Strains at Field-A

A45S2

(b) Lower Level Ground Strains at Fleld-A

B2 CO

B96C0

(c) Lower Level Ground Strains at Fleld-B

Figure 6 Tetrahedrons Foraulated to Calculate Ground Strains

175
OOOO OC'OO oooo oooo oooo oooo

i

176
5 7 10 20 30 SO 70 100 300 )00 500 700 1000 EPICENTBOL 0I5TRNCE A IKMI

20 3D 50 70 100 200 300 500 700 1000 EPICENIROL OISTflNCE A IKHI

(1) MaxlBUB Noraal Strain e (2) Maxlaua Shear Strain 7 (a) Upper Level Ground Strains at Field A

LOWER LEVEL A-FIELO I M |l

e 30

0 4SM<5

SSM<6

esM<7

7SM

I

I

S 7 10 ;0 30 so 70 100 JOG 300 SOO TOO 1000 EPICEMIRPI, 0I5TBNCE a IKMl

20 30 SO 70 100 ;00 300 SOO 7 00 1000

(1) MaxlauM Noraal Strain e (2) Naxlaua Shear Strain 7

(b) Lower Level Ground Strains at Field A

3 3 S 7 10 - 30 30 50 70 lOO 300 300 500 700 1000 EPICEnrRfn. OfSTPUCE 6 ikhi

5 7 10 20 30 50 70 100 2C EPrCENTRRL OlSTFlNCe d IKMI (1) MaxlHua Noraal Strain e (2) Haxlaua Shear Strain 7

(c) Lower Level Ground Strains at Field B

Figure 8 Attenuation Characteristics of Maxlaua Ground Strains

177
Dynamic Centrifugal Model Tests of Embankments

on Liquefiable Grounds

BY

Yasuyuki Koga' and Jun-ichi Koseki^

ABSTRACT

Model tests using a geotechnical dynamic

centrifuge were performed to study a seismic

behavior of embankments on liquefiable grounds.

Effects of a scale factor and a seismic history

were investigated. Tests on a shape of the embankment or a thickness of the ground were also

conducted

.

KEY WORDS : Embankment, Geotechnical Centrifuge,

Liquefaction

A sinusoidal shaking was conducted after

applying a centrifugal acceleration. It was repeated several times increasing an amplitude. A summary of the test cases and the test conditions

are shown in Table 1. A prototype wave frequency

in the Ig field was 2Hz for both the 30g models

and the 50g models. An effect of a seismic

history was investigated by comparing the results

obtained in the first shaking steps with those

obtained after them.

3. EFFECT OF SCALE FACTOR

1. INTRODUCTION

Liquefaction of a sandy ground causes a settlement of an embankment founded on the ground. A lot of model tests have been executed

to reveal the seismic behavior of the embankment,

but few of them have been done under a high

confining stress.

A centrifugal model test is a test method which

can simulate the same stress condition in a scaled model ground as in an actual ground. It has been applied to a dynamic problem using an earthquake simulator" . This paper deals with dynamic centrifugal model

tests of a horizontal sand layer and an embankment on the layer.

2. TEST PROCEDURE

Fig.l shows a cross section of test models.

They are classified into horizontal layer models

and embankment models. An effect of a scale

factor was investigated using 30^ models and 50^ models. The model dimensions were determined for both models so that a prototype size in a Ig

field may almost coincide with each other. A shape of the embankment and a thickness of the sand layer were changed in models I-l and 1-2.

A sand layer was prepared by pouring Toyoura

sand through air in a rigid soil container. The

surface of the layer was rounded in accordance

with a rotational radius. In the embankment

models a colored sand mesh was drawn to observe

the deformation of the sand layer through a transparent front glass of the container.

After setting the container in a vacuum box the sand layer was saturated with silicone oil which

is 30 times as viscous as water for the 30^ models and 50 times for the 50g models. An embankment was made of a mixture of Toyoura sand

and a clay-sand which has a ratio of 4:1 in weight and a water content of 15%. Fig. 2 shows a location of transducers.

3 . 1 Comparison of Time Histories

Fig. 3 shows measured data of the horizontal

layer models, and Fig. 4 shows those of the embankment models. Time axes in the figures are controlled to have the same prototype length in the Ig field for the 20g model and the 50£f model.

In Fig. 5 and Fig. 6 the time axes are shortened to show changes after the shaking.

There were qualitative agreements between the result of the 30sr models and that of the 50g

models as fol lows:

(1) An increase of an excess pore pressure

proceeded faster in a shallow part of the horizontal sand layer than in a deep part, and a rapid change of a responce acceleration occured

earlier in the shallow part. There was almost no acceleration responce after the liquefaction. The excess pore pressure dissipated faster in the deep part.

(2) A change of an excess pore pressure beneath

the embankment (P4 in the embankment model) was small during the shaking, but it increased after

the shaking. An excess pore pressure dissipated

faster below the embankment (P6) than in the side

layer (P3) of the same depth.

(3) The rapid change of a responce acceleration

did not occur in or below the embankment (A4 and

A6) while it occured in the side layer (A2).

(4) A crest settlement of the embankment started

as the excess pore pressure increased. It proceeded at a nearly constant rate during the shaking, but it stopped after the shaking.

The dissipation of the excess pore pressure

finished in about Ssecond in the 30g models, and

it did in about Ssecond in the 50g models. The

duration for both models was equivalent to ISOsecond in the Ig field. It increased as the excess pore pressure generated fast.

'Head, Soil Dynamics Division, Construction

Equipment and Method Department, Public Works

Research Institute, Ministry of Construction,

Japan. ^Researcher, ditto.

178
In this test the viscosity of the silicone oil used as a pore fluid of a sand was determined

assuming that it is in inverse proportion to an apparent permeability of the sand although

Inatomi et al.^' showed experimentally that it is not so. Within the limits of this test it may

well be adequate to use the simple assumption,

because the duration of the dissipation was almost equivalent.

3.2 Resistance against Liquefaction ' Fig. 7 shows the relationship between the input

acceleration and the number of cycles until the rapid change of the response acceleration (A4 to A6) due to the liquefaction in the first shaking

steps of the horizontal layer models. Fig. 8 shows

one between the input acceleration and the number

of cycles until the excess pore pressure reaches

a limiting value which is maintained during the liquefaction. The input acceleration is converted

into a prototype value in the Ig field to compare

the result of the 30g models with that of the 50g

models

.

It is seen from the figures that the resistance

of the sand layer against a liquefaction was

lower in the 50g models than in the 30g models.

Although the reason is not known yet, it is possible that a change of a frictional resistance

at a contact point of the sand particles may have

reduced the resistance of the sand in an undrained condition as the viscosity of the silicone oil increased.

3.3 Crest Settlement

Fig. 9 shows the relationship between the input

acceleration and the induced crest settlement of the embankment in the first shaking steps of the embankment models. Both values are converted into

prototype ones in the Ig field. The prototype

settlement was larger in the 50g models than in the 30g models if the prototype input

acceleration was the same. It is estimated that

the difference is caused by the change of the resistance against the liquefaction, so the effect is investigated as follows:

(1) Fig. 10 shows the relationship between the prototype input acceleration and the number of cycles until the responce acceleration in the side layer (A2) changes rapidly due to the liquefaction. The 503 models had a lower

resistance against the liquefaction than the 30g

models as is seen in the Fig. 7. (2) Fig. 11 shows the relationship between the prototype crest settlement and the number of cycles used in the Fig. 10. There was almost no difference between the 20g models and the 50g

models. This number of cycles can be assumed to represent the resistance of the whole layer

against the liquefaction, then it is estimated

that the prototype crest settlement in the 50g

models may be almost the same as in the 30g

models if they have the same resistance against

the liquefaction.

(3) A prototype crest settlement in several

shaking steps with a seismic history is compared

between the 30^ models and the 50g models in Fig. 12, and the number of cycles defined as in the Fig. 10 is compared in Fig. 13. Both the induced settlement and the accumulated settlement

were larger in the 5Qg models than in the 30£r models, and the 50g models had a lower resistance

against the liquefaction. The results agree with

those obtained in the first shaking steps.

4. EFFECT OF SEISMIC HISTORY

The effect of a seismic history was investigated on the result obtained in the six

cases of model 63-1. The shaking amplitude was changed in the first step of each case, and it was increased gradually in the following steps.

4.1 Crest Settlement

Fig. 14 shows the relationship between an input

acceleration and an induced or accumulated crest settlement. The induced settlement in the first

step of each case without the seismic history was larger than the one in the following steps of cases 63-1-2 and 63-1-3 with the seismic history,

and it was smaller than the accumulated

settlement of the pre-shaken cases. The difference in the accumulated settelement

was small between these cases, therefore the case 63-1-2 is used as a representative case with the seismic history in the following comparisons.

4.2 Excess Pore Pressure

Fig. 15 shows the relationship between the input

acceleration and a maximum excess pore pressure

ratio at a depth of 10cm (P2 and P5). An initial

effective vertical stress is calculated

one-dimensional ly using a measured unit weight of the sand layer and the embankment.

With or without the seismic history the ratio

in the side layer (P2) was nearly 1.0 when the input acceleration was about Ag , but the ratio was higher with the seismic history when the input acceleration was more than 7g. Below the embankment (P5) the ratio was higher with the seismic history than without it over a whole range of the input acceleration.

It is considered that with the seismic history

a densi f ication of the sand layer due to a reconsol idation after the liquefaction reduces

the generation of the excess pore pressure, and

the rate of the dens i f i cat i on seems to be larger

below the embankment than in the side layer.

4 . 3 Response Acceleration

Fig. 16 shows a relationship between the input

acceleration and a ratio of a maximum response acceleration to the input one. With or without

the seimic history the ratio was almost the same

in the side layer (A2). Below the embankment (A6)

179
it was higher with the seismic history when the input acceleration was more than Ig. In the embankment (A4) the ratio increased as the input

acceleration increased with the seismic history,

and it decreased without the seismic history.

These results agree with the consideration that the effect of the dens i f i cat i on due to the seimic

history is higher below the embankment than in the side layer. The difference in the induced crest settlement between the virgin cases and the pre-shaken cases may be caused by both the disification and the residual deformation of the pre-shaken cases.

5 EFFECT OF EMBANKMENT SHAPE AND SAND LAYER THICKNESS

5 . 1 Crest Settlement

A crest settlement induced in the first steps

is compared between models 63-1, 1-1 and 1-2 in Fig. 17. There was not a distinct difference, but

the settlement was larger in models 1-1 and 1-2 than in model 63-1 when the input acceleration

was about bg.

Generally it is estimated that the thinner sand

layer in model 1-1 may result in the smaller

settlement than in model 63-1, and that the larger embankment in model 1-2 may result in the larger settlement than in model 63-1. Altouogh

the reason for the disagreement of the test results with the estimation is not known yet, it is possible that a small change in the model

preparation may have caused variations in the characteristics of the models such as the density

and the saturation degree of the sand layer which

affected the test results. Further investigations

are needed to evaluate the effect of these

variations, and a careful preparation of the model with a high repeatability is necessary.

5 . 2 Deformation of Sand Layer

Fig. 18 shows an observed deformation of models

63-1, 1-1 and 1-2. A settlement of the embankment

accompanied with a lateral deformation of the sand layer and a settlement of the layer below

the embankment. The lateral deformation was larger in a shallow part of the layer, and it was restrained by a bottom plate of the container.

The settlement of the layer may have been caused

by both the lateral deformation and the reconsol idation after the liquefaction. There

could not be seen a distinct difference among the three models.

6 CONCLUSIONS

The result of dynamic centrifugal tests performed using horizontal layer models and embankment models with a centrifugal acceleration

of 30^ or 5O3 is summarized as follows:

(1) A comparison of the measured data of the 20g

models and the bOg models showed that there were

several qualitative agreements between them on a dynamic response characteristics. A lower

resistance against a liquefaction in the 50g

models resulted in a larger prototype crest settlement, and it may have been caused by the change of a viscosity of a silicone oil used as a pore fluid. It is estimated that the prototype

crest settlement in the 50g models may be almost

the same as in the 20g models if they have the same resistance against the liquefaction.

(2) To investigate the effect of a seismic

history a comparison was made between the results

obtained in the first shaking steps of the embankment models and those obtained in the following steps. The effect on the excess pore pressure and the response acceleration was larger

in or below the embankment than in the side

layer. It is considered that the rate of a dens i f i cat i on of the sand layer due to a reconsol idation after the liquefaction is larger

below the embankment than in the side layer. A smaller induced crest settlement with the seismic

history was obtained, and it may have been caused

by both the dens i f i cat i on and the residual

deformation.

(3) The test result did not clarify the effect of

a shape of an embankment and a thickness of a sand layer on the crest settlement. The

settlement of the embankment accompanied with a lateral deformation of the sand layer and a settlement of the layer below the embankment

irrespective of the shape or the thickness.

REFERENCES

1) Koga, Y. , Taniguchi, E. , Koseki , J. and Morishita, T : Sand Liquefaction Tests Using a Geotechnical Dynsamic Centrifuge, Proc. of the 20th Joint Meeting of U.S. -JAPAN Panel on Wind

and Seismic Effects, US/JAPAN Natural Resources Conference, 1988.

2) Inatomi, T. , Kazama, K. , Noda, S. and Tsuchida, H. : Centrifuge Dynamic Model Tests in PHRI, Proc. of the 21th Joint Meeting of U.S. -JAPAN Panel on Wind and Seismic Effects,

US/JAPAN Natural Resources Conference, 1989.

3) Ledbetter, R. H. : Modelling of Earthquake

Induced Pore Pressure Behavior, Proc. of the 21th

Joint Meeting of U.S. -JAPAN Panel on Wind and Seismic Effects, US/JAPAN Natural Resources Conference, 1989.

180
Table 1 Test Cases and Conditions

Model Case

Cen tr i f ugal Acc , ig)

Input Sinusoidal Wave

Input Acc. ig) Dr of Sand

1st 2nd 3rd 4th 5th 6th Layer (%)

63-1

63-1-1

63-1-2

63-1-3

63-1-4

63-1-5

63-1-6

30 60Hz,20cycle

1.2 2.5 3.7

5.5

7.3 9.0

4.1 5.6

5.2

7.1

7.3

8.3

8.7

12.5

12.7 *

**

*

1-1 1-1-1

1-1-2 30 60Hz,20cycle

3.4

4.9

4.9

6.3

6.3 8.08.0 61

62

1-2 1-2-1 1-2-2 30 60Hz,20cycle

3.3 5.05.3 6.7

7.0

8.5 8.8 6470

1-3

1-3-1 1-3-2

1-3-3

30 60Hz,50cycle

2.3 3.4

4.5

3.3

4.3 5.6

4.5

5.2

5.6 64

6159

1-4

1-4-1 1-4-2 1-4-3

50 100Hz,50cycle

4.0

4.8

2.6

5.3

6.4 3.8

7.0

7.8

8.4 64

6164 1-5 1-5-3

1-5-4

50 100Hz,20cycle

7.0

5.8

10.8

9.2

14.0

11.7

16.1 13.9

5260

* The relative density of the sand layer for model 63-1 is not known due to mismeasurement

.

It is assumed to be almost the same as others, because the sand layer was prepared in the same way.

!

I

. „. , , _ . : Accelerometer (Horizontal)

i- D.sp acement Transducer ^, Accelerometer (Vertical) R: Rotational Radius q: po,e Pressure Transducer

Unit:/jun

Sic

V

53 J_

(1) Model 63-1 (5) Model 1-4

63 36 63

o »J o

.

ID O

in - LD

Q. "Q, OLD "o

30g model

50g model

AI OPI OA2 oP2

A3 0P3

(2) Model 1-1

'.00 250 AV^AO

(250) (150)

-I 150

(100)

(1) Models 63-1,1-2,1-5

(1) ~ (4) : 30g models

(100mm in width)

(5) , (6) : bOg models

(60mm in width)

(3) Model 1-2

800

(4) Model 1-3

Fig.l Test Models

A6 oP5

R = 2300 OA7

A3 OP3 °P1 M

OPA °P2

cu

/.OO 150 100 150

(2) Model 1-1

p_ R = 2100(2030)

A4 30g model ^ Sas

50£r model \ oaI

R = 2300 oP6

^^(2150K

DAI opi

A2 oP2

a A3 oP3

/.OO 250 AV°"AO

(250) (150)

(3) Models 1-3,1-4

(100)

Fig. 2 Location of Transducers

181
20 rTT-

Acceleration i ,„f ^ AO (g)

0

-20

Excess Pore 1.0 Pressure pa (kgf/cm^)

P5 (kgf/cm^)

P6 (kgf/cm2)

0 Time (msec) 5000 10000

(1) Case 1-3-2 (30g) in the 1st shaking

20

0

-20

1.0

0

1.0

0

1.0

0

0 Time (msec) 3000 6000

(2) Case 1-4-2 (50g) in the 1st shaking

Fig. 5 Measured Data of Horizontal Layer Models (Changes after the shaking)

20

Acceleration

AO (g) 0

-20

25

Settlement D V (mm) 0

Excess Pore Pressure P3 (kgf/cm^)

0

1.0

0

1.0

shaki ng

1.0

P 4 (kgf/cm^)

P 6 (kgf/cm^)

0

0 Time (msec) 5000 10000

(1) Case 63-1-3 (30g) in the 1st shaking

0 Time (msec) 3000 6000

(2) Case 1-5-4 (50g) in the 1st shaking

Fig. 6 Measured Data of Embankment Models (Changes after the shaking)

200

« 5 150 U cn V

S I 100 3 -a a —

c <v 50

0

in the 1st sliaking Model A 4 A 5 A 6

1-3 (30g) 0 A

1-4 (50g) •

II C —

o o — c» o

03 00

200

150

S i 100

50

0

0- <= 0.5 1 10 20 50 0.5 2 10 20 50

Number of Cycles Until Rapid Change

of Responce Acceleration

Fig. 7 Input Acc. vs. Number of Cycles

Until Rapid Change of Responce Acc. (Horizontal layer model)

Number of Cycles Until Excess Pore Pressure Reaches Liquified Value

Fig. 8 Input Acc. vs. Number of Cycles Until Excess Pore Pressure Reaches Liquified Value

(Horizontal layer model)

182
20

Acceleration

AO (g)

-20

Excess Pore 1.0 Pressure P4 (kgf/cm2)

0

P 5 (kgf/cm2)

P6 (kgf/cm=)

0 L_jyVVV\AAAA/VVVWVW*AAAyWV\AAAMAA^ —

Acceleration

A 4 (g) 0 h-vvwvyiV^

-20

20

20

0

-20

1.0

0

1.0

0

1.0

A5 (g) °

A 6 (g)

-20

20

0

-20

VNAA/VVVWVVVVVVVWVVVVWVW/fV-'

0 Tine (msec) 500 1000

(1) Case 1-3-2 (30g) in the 1st shaking

0 Time (msec) 300 600

(2) Case 1-4-2 (50g) in the 1st shaking

Fig. 3 Measured Data of Horizontal Layer Models

20

Acceleration

A 0 (g) 0

-20

25

Settlement D V (mm)

WVWWWVWVWWVWV—

20 [ 0

Excess Pore Pressure P3 (kgf/cm^)

P a (kgf/cm^)

-20 L—

25

P6 (kgf/cm2)

Acceleration

A 2 (g)

A 4 (g)

0

20

0

-2020

-2020

A6 (g) 0

-20

0 Time (msec) 500 1000

(1) Case 63-1-3 (30g) in the 1st shaking

0

1.0

0

1.0

0

1.0

0

20

0

-20

20

0

-20

20

0

-20

0 Time (msec) 300 600

(2) Case 1-5-4 (50g) in the 1st shaking

Fig. 4 Measured Data of Embankment Models

183
U (V

50

AO

30

20

10

0

1

A :

—11 1 1 ]

63-1-3 ^ V: 63-1-2 \nOg)

<• 63-1-1-6-'

in the 1st shaking

- ^.^^^-"""^'^

1 V r 1 > 1 1 0 50 100 150 200 250 300

Prototype Input Acceleration

in the Ig field igal) , 980gal = lg

Fig. 9 Input Acc. vs. Crest Settlement

350

' AOO

c

•- §> « § 300

o i 200

100

1 1 1 A :

r - 1 1 63-1-3 -^ V : 63-1-2 >(3O0) < < : 63-l-1~6-' < < i n the 1st shaking

A

1 ' 1 1 0.5 1 2 5 10 20

^ Number of Cycles Until Rapid Change

= of Responce Acceleration A 2

Fig. 10 Input Acc. vs. Number of Cycles Until

Rapid Change of Responce Acc. A 2

AO

30

20

10

0

1

1 1 A :

1 1 1 63-1-3 ^ V: 63-1-2 ){30g)

<: 63-1-1-6-' <

:

: < <

i n the 1st shaking

<

1 [__

A

0.5 1 2 5 10 20

Number of Cycles Unti I Rapid Change

4) S

CO u

<u —

.

0 —

M-l

1 ^

o

70

60

50

^0

30

20

10

0

(30g) induced/Accumulated

( 63-1-3 : -'--^ / ^

i 63-1-2 -^-y-

(50g)

1-5-3 -o-/-*-- /

1-5-2 :-^ -/--^- -• ^

in all shakings . / /'

0 50 100 150 200 250 300 350 Prototype Input Acceleration

in the Ig field igal), 980gal-lg

Fig. 12 Input Acc. vs. Crest Settlement

(wi th seismic history)

.5 AOO

.2 °

« I 300

200 -

g-? 100

S ^ 0

in all shakings a:63-1-3

v: 63-1-2 X

1-5-3

(30fir)

}(50sf)

v-»

05 1 2 5 10 20

Number of Cycles Until Rapid Change

; of Responce Acceleration A 2

Fig. 13 Input Acc. vs. Number of Cycles Until

Rapid Change of Responce Acc. A 2

(with seismic h i s t ory

)

of Responce Acceleration A 2

Fig. 11 Crest Settlement vs. Number of Cycles Until

Rapid Change of Responce Acc. A 2

Notation for Fig. 14 Seismic History (Case) Type of Settlement

1 nduced Accumulated

V i rgi n (&3-1-1 ~6)

Pre-shaked

(63-1-2) - -

Pre-shaked

(63-1-3)

2 A 6 8 10 Input Acceleration (.g)

Fig. 14 Input Acc. vs. Crest Settlement (Model 63-1)

184
Notation for Fig. 16 Notation for Fig. 15 Seismic History (Case)

Loca t ion P 2 P 5

Virgin

(63-1-1 ~6) •

Pre-shaked

(03-1-2) A- - - --

0 2 4 6 8 10 Input Acceleration (.g)

Fig. 15 Input Acc. vs. Excess Pore Pressure Ratio (Model 63-1)

16U

12

10

I I

- -0-- Model

1 I

63-1

Model 1-1 -a— Model 1-2

in the 1st shaki ng

Y t U 1 J 1 / I / 1 J A

1 1 1 1

0 2 6 8 10 Input Acceleration (g)

Fig. 17 Input Acc. vs. Crest

Se 1 1 1 emen t

o

<

s <

Seismic His- tory (Case) Loca t i on A 2 A 4 A 6

V i rgi n (63-1-1 ~6) —o

—•

Pre-shal^ed (63-1-2) - - -A - --^ _ _ _4_ _ _

0 2 4 6 8 10 Input Acceleration, AO {g)

(1) In the side layer

0 2 4 6 8 10 Input Acceleration, AO (5)

(2) In or below the embankment

Fig. 16 Input Acc. vs. Response Acc. Ratio (Model 63-1)

I ' lOOmm

rbefore the 1st shaking

:after the 3rd shaki ng

rbefore the 1st rafter the 5th

(1) Case 63-1-3

- rbefore the 1st — rafter the 3rd

(2) Case 1-1-2

rbefore the 1st rafter the 3rd

(3) Case 1-2-2

Fig. 18 Observed Deformation

185
Seismic Resistance Requirements of Roller Compacted Dams

by

Alan T. Richardson^ and Louis H. Roehm^

ABSTRACT

Seismic analysis of the U.S. Bureau of

Reclamation's (Reclamation) Upper Stillwater

Dam is explained as an example of a design

method used for evaluation of an RCC dam's

seismic resistance. Material strength

properties were evaluated by laboratory and

field testing. Testing evaluated the weakest

link in an RCC structure, the joint between

lifts. Computer code EADHI was used to analyze finite element cross-sectional models

of the dam for static and dynamic loads. The

maximum credible earthquake (MCE) was scaled

by spectrum intensity to determine the site

response spectrum. Synthetic accelerograms

were generated for horizontal and vertical

ground motions from the site response

spectrum.

Maximum tensile and compressive stresses were determined from the dynamic analysis.

Compressive stresses were well below

acceptable limits for the RCC. Tensile

stresses on the upstream face exceeded the

allowable strength of the RCC for 0.10

seconds, indicating cracking of the concrete

could occur.

1. INTRODUCTION

Roller compacted concrete (RCC) or a roller

compacted dam (RCD), are terms which refer to

a construction method for placing mass concrete with large equipment normally used

for hauling rock or soil, and compacting the

material in place with large vibrating

rollers. Time and cost savings achieved by

this procedure make it a preferred

construction method for concrete dams.

Seismic resistance of a dam for a given

seismic acceleration depends primarily on the

dam's shape, material strength and the stress

state existing in the dam. The dam's shape

is typically determined by an economic

analysis of structurally stable cross sections based on static loads, with shape

considerations given for dams located in areas of higher seismicity. Specific

material properties for a dam are determined

by standard laboratory testing and adjusted

for dynamic loads by applying empirically

determined factors based on rapid strain rate

testing. For RCC dams, material properties

testing also requires careful evaluation of

the bond strength to be developed between » lift lines, since bond strength generally

controls the dam's ability to resist vertical

tensile stresses. The stress state existing

in the dam prior to the dynamic analysis is determined by analysis of static loads on the

dam. Dynamic analysis is then used to determine the dam's response to seismic

loads.

11! Upper Stillwater Dam was designed in 1982 by

Reclamation. Its design is used here to demonstrate the analysis and evaluation

process needed to determine an RCC dam's

seismic resistance. Upper Stillwater Dam

(figure 1), located in the state of Utah at an elevation of 8,000 feet, was constructed

between 1983 and 1987. It has a crest length \\ of 2,700 feet, a maximum structural height of

285 feet and contains a total of 1,600,000

cubic yards of concrete, 90 percent of this

total is RCC. Construction was by placing

and compacting 1-foot (30 cm) thick lifts

between the abutments until the total height

of the dam was placed. Both faces of the dam were formed by continuous horizontal

slipforming of very low slump conventional

concrete. Contraction joints were formed

naturally as the dam cooled. The selected

dam's cross section minimized the concrete

volume while providing the required stability

safety factors for static loads. It was then

analyzed to determine its ability to resist

seismic loads. The dynamic analysis of Upper

Stillwater Dam was conducted using the finite

element program EADHI [A]. In this program,

the complete system is considered as composed

of two substructures, the dam represented as

a finite element system, and the water in the

reservoir treated as a continuum of infinite

length in the upstream direction governed by

the wave equation. Compressibility of the

water is included in the analysis. EADHI is

a two-dimensional program including both

horizontal and vertical components of ground

motion. The program uses an isoparametric

^ Principal Designer,

U.S. Bureau of Reclamation,

Denver, Colorado 80225

^Technical Specialist,

U.S. Bureau of Reclamation,

Denver Colorado 80225

186
quadrilateral element. The material behavior

is assumed to be linearly elastic.

2. MATERIAL PROPERTIES

Reclamation's concrete laboratory conducted

an extensive testing program to determine the

material properties of RCC [B]. This program

included standard structural properties

testing of laboratory cylinders and core from

a test placement constructed in 1981 at the

dam site. Structural testing emphasized the

tensile capacity of the joint created between

layers of the RCC. Direct tensile tests were conducted on jointed cylinders and at the

joint between lifts in cores.

Dynamic properties were not obtained in the

laboratory, so static values were increased

proportionally according to percentages

reported by others [C]. The static

compressive strengths of the concrete were increased 20 percent to account for rapid

j loading effects. This assumption resulted in

f ultimate dynamic compressive strength of

' 3,600 lb/in2 for RCC. The static tensile

! strength was increased 60 percent to account

for rapid loading effects. In addition, the

tensile strengths were increased 20 percent

j to account for nonlinear behavior of the

; concrete. The analytical models used assume

the concrete to be elastic when actually it

'] behaves nonlinearly at failure. Since the

I stresses computed are therefore artificially

I high, the assumed tensile strength should be

increased accordingly. The ultimate dynamic

il tensile strength for RCC resulting from these

assumptions was 350 Ib/in^ . The dynamic

i modulus of elasti-city for all concrete was assumed to be 2.5 (lO*) Ib/in^, which is 67

I percent greater than the static value. This

• increase is documented by laboratory tests

performed by others [C]. The dynamic modulus

j of deformation of the foundation rock was

! assumed to be 1.25 (lO"^) Ib/in^. This value

ij was based on analytical studies of modulus of

I 1.0 (lO^) lb/in2 to be similar to results

' from studies with several materials in the

I foundation. Increasing the modulus for

I dynamic loading resulted in a value of

I 1.25 (106) ib/in2.

In addition to material properties, the

' dynamic analyses required a viscous damping

I, constant. Several experimental shaking tests

[j on models and prototypes of concrete dams

i have indicated values ranging from 2 to

10 percent [D]. The smaller damping values

were associated with very low levels of

I excitation. For earthquake-type vibrations,

I a value of 10 percent is more representaI tive. For the dynamic analyses of Upper

Stillwater Dam, damping was assumed to be

10 percent of critical.

3. INPUT GROUND MOTIONS

In the Reclamation process, two steps are necessary in selecting the seismic loadings

to which a dam may be subjected. The first

step is to define the magnitudes and

locations of the hypothetical earthquakes

which could affect the site. The second step

is to determine the response spectra and

select accelerograms representative of the

resulting rock motions.

The earthquake magnitudes and locations

estimated for the Upper Stillwater site are listed in the following table.

Richter magnitude

MCE 100-vr 25-yr

Epicentral

di stance

Focal

depth

6.0 5.2 4.3 2 km 7 km

The dam was analyzed only for the maximum

credible earthquake (MCE) since it would

cover the 100- and 25-year events.

Ground motions for use in dynamic analyses were determined as follows: First,

historical records with magnitudes and

locations similar to the MCE were scaled

using spectrum intensity. Spectrum intensity

is defined as the area under a velocity

response spectrum between 0.1 and 2.5

seconds. Then the average spectral

accelerations were calculated for each period

of the scaled records. The resulting site

response spectrum (SRS) is shown in figure 2.

Once the SRS was determined, synthetic

accelerograms were generated. Independent

accelerograms were generated for the

horizontal and vertical components of the

ground motion. The vertical record was computed by scaling one accelerogram by the

ratio (0.48) of vertical to horizontal

spectrum intensities of the historical

records. The horizontal and vertical

accelerograms are shown in figures 3 and 4,

respectively. Since the Upper Stillwater

analysis. Reclamation has refined the

spectrum intensity concept (reference E).

Computations and field tests have

demonstrated that the natural periods of

concrete dams are generally less than

0.5 second. Therefore, acceleration spectrum

intensity has been introduced and defined as the area under the accelera-tion response

spectrum between periods of 0.1 to

187
0.5 seconds and is an appropriate indicator

of the potential response of concrete dams to

a given earthquake record.

4. DYNAMIC ANALYSIS

Dynamic studies of Upper Stillwater Dam used

the two-dimensional program EADi'I. Two

models were used in the analyses - one to represent the maximum section and a more

typical height section. Hydrostatic loads

analyzed were: the normal water surface,

elevation 8172.0, and the minimum water

surface, elevation 8028.4.

Model grids used in the dynamic studies are shown in figures 5 and 5. Smaller elements were used at elevation 8100, the reentrant corner on the downstream face, to provide

better stress definition.

All dynamic studies used only one material

model in the dam. The foundation elements

used another massless material for rock.

Studies made with a homogeneous foundation

showed tensile stresses exceeding the

capacity of the rock in the foundation

element just upstream of the dam. In order

to provide a more realistic model, this

element was softened by lowering the

deformation modulus of the element to 100,000 lb/in2.

Three different stress output options were

used in the EADHI analyses. The first was

used to obtain a table of maximum and minimum

stresses, and their respective times of occurrence for each element. Secondly,

stress histories were generated for

particular elements. The third option was

used to obtain the stress distributions in

the entire dam at specific times

corresponding to maximum tensions on both

faces. EADHI automatically added static

stresses to the dynamic stresses in all

analyses.

Several studies were made to determine how

much influence individual modes had upon the

response of the dam. Up to five modes were included in each study.

5. RESULTS

Each EADHI analysis produced a large amount

of output. Only the most significant results

are presented. Modal superposition analyses

were conducted which included contributions

of up to five modes. The following results

refer to studies for the first four modes;

the fifth mode was judged to have a negligible influence upon the response of the

dam. Natural periods for the maximum and

typical sections are listed in the following

tabl e:

Natural periods - seconds

Maximum Typical

Mode section section

1 0.462 0.344

2 0.182 0.139

3 0.176 0.125

4 0.094 0.063

The maximum principal stresses orcurring at each face are shown on figures 7 and 8.

The maximum tensile stresses of 432 Ib/in^ on the upstream face and about 300 Ib/in^ on the

downstream face occur when the water surface

is at elevation 8172 (figure 7). For the low

water conditions, maximum tensions are about

200 Ib/in^ on each face. At the maximum

section, the maximum compressive stress (not

shown) is 1,042 Ib/in^. Maximum tensions at the upstream face of the typical section

with the normal water surface are about

300 Ib/in^. Vertical stress distributions in

the maximum section at particular times

related to the maximum tensions on both faces

are shown on figure 9. Maximum tensions of

about 300 Ib/in"^ occur on the upstream face

at elevation 8040 with normal and minimum

water surface elevations at times 5.19 and

2.23 seconds, respectively, in the response

history analyses. Maximum tensions of about

200 Ib/in^ occur on the downstream face at elevation 8095 at times 4.32 and 3.05 seconds

for normal and minimum water surface

elevations, respectively. As can be seen in

figure 9, the stress distribution is almost

linear from face to face.

In order to get an indication of the duration

and frequency of occurrence of the maximum

tensile stresses, several studies were conducted to generate stress histories at selected elements in the maximum section.

These elements are 132, which is situated at elevation 8040 on the upstream face, and 167,

located at elevation 8095 on the downstream

face. Stress histories for element 132 for

the analyses including normal and minimum

water surface elevations, respectively, are shown on figures 10 and 11. Corresponding

stress histories for element 167 are included

in figures 12 and 13. Tensile stresses

exceed 250 Ib/in^ only in the element 132 for

the normal water surface condition; all four

188
'] such excursions are of durations less than

0.10 second each.

|!| All dynamic analysis results indicate stress

' levels in the dam for the minimum water

condition are significantly lower than for

,1 the normal water surface loading. In

''I addition, stresses from the analyses of the

typical section are substantially reduced

compared to stresses from the maximum section

analysis.

6. EVALUATION

According to Reclamation criteria [F], a safety factor greater than 1.0 is required

for the extreme loading combination. The

allowable compressive stress for the RCC is

' 3,600 Ib/in^. The allowable tensile stress

for RCC is 350 Ib/in^.

Response history analyses of the normal water

condition indicate maximum compressive

. stresses in the RCC less than 1,100 Ib/in^

I which are well below the allowable limit of

3,600 Ib/in^. Tensile stresses in the RCC

I exceed the allowable limit of 350 Ib/in^ for

less than 0.10 second at the upstream face,

i' Except for four excursions of short duration,

1 tensile stresses are generally less than

I 250 Ib/in^ throughout the dam. When large

tensile stresses are present at the upstream

' face, the downstream half of the dam is in

compression. Also, the above-mentioned

h stresses refer to the maximum section

''i analyses; stresses at the typical section are substantially lower.

|1 Present Reclamation criteria state that

"Horizontal cracking should be assumed to occur in a gravity dam wherever the vertical

li normal stress for dynamic response to an

i' earthquake does not meet minimum safety

factor requirements. The depth of crack is

assumed to extend along a horizontal section

to the point where compressive stress

|| computed without uplift and the internal

|j hydrostatic pressure are equal." These

' criteria, published in 1974, are based on a

' pseudo static analysis for earthquake

I loadings. This method does not account for

I the rapid cyclic loading of seismic

excitation. The analysis method used for

this design (EADHI) produces a stress history

based on linear elastic theory while

accounting for resonance effects in the dam

and hydrodynamic interaction. In addition,

earthquake-type ground motions are input to the analysis rather than applying peak

' acceleration values to the concrete and water

1 mass. Consequently, the evaluation of

1 results from the EADHI program requires a different approach than outlined in E.M. 19.

It is important to note that the dynamic

tensile strength was exceeded for a short

duration on the upstream face. An excursion

beyond the elastic limit is not as significant as the general level of

repeatable stress at a point. An examination

of the stress histories for this analysis

shows that the dam responds at repeatable

stress levels safely below the dynamic

tensile strength. On this basis, it is

reasonable to conclude that cracking would

not develop through the section. From these

observations, the dam's response is predominantly linearly elastic with instants

where localized portions of the dam become

nonlinear. Therefore, the loading from the

MCE could possibly cause cracking of the

concrete in a limited area. However, the dam

possesses sufficient stability to operate

safely during and after the earthquake.

One method of substantiating judgements

concerning structural behavior is to observe

the actual behavior of similar prototype

structures for similar loading conditions.

In 1967, Koyna Dam, a 338-foot-high gravity

dam located in India, was shaken by a Ms = 6.5 earthquake located at an epicentral

distance of 5 miles and a fault break

distance of 1.8 miles. The structural damage

to the dam was horizontal cracks on either or both faces of a number of monoliths [G, H].

Leakage of water was observed on the

downstream face of one monolith, and traces

of seepage were observed on five other

monoliths. Although the dam cracked, no sudden release of the reservoir took place.

The Koyna Dam experience shows that a concrete gravity dam like Upper Stillwater

can crack and still retain the reservoir.

Stresses from the low water analyses were substantially reduced compared to stresses of

the normal water condition analyses.

In summary, the evaluation of the dynamic

analysis of Upper Stillwater Dam indicates

that the maximum credible earthquake could cause cracking, but no sudden release of the

reservoir will occur. This condition is permissible within the present Reclamation

policy.

In August 1984, the Earthquake Engineering

Research Center released the program EAGD-84

[I]. This program supersedes EADHI and

includes the effects of dam-water-foundation

rock interaction and of materials, such as alluvium and sediments, at the bottom of

reservoirs. EAGD-84 was run using the Upper

Stillwater data. The results indicated that

stress values at the heel of the dam were

189
very similar and maximum values were 100 to 150 Ib/in^ less in the upper part of the dam.

This program has become the standard for two- dimensional gravity dam analysis at Reel amation

.

9. References

[I] Chopra, A. K., and G. Fenves, "EAGD-84 - A Computer Program for

Earthquake Analysis of Concrete Gravity

Dams," Report No. EERC 84-11, College of

Engineering, University of California,

Berkeley, 1984.

[A] Chakrabarti, P., and A. K. Chopra,

"EADHI - A Computer Program for

Earthquake Analysis of Gravity Dams

Including Hydrodynamic Interaction,"

Report No. EERC 73-7, College of

Engineering, University of California,

Berkeley, 1973.

[B] Memorandums from Chief, Concrete and

Structural Branch, to Chief, Dams

Branch, "Upper Stillwater Dam - Roller

Compacted Concrete," April 7, 1982, June

24, 1982, August 30, 1982, and January

13, 1983.

[C] Raphael, J. M., "Tensile Strength of

Concrete," Journal of the American

Concrete Institute, March-April, 1984.

[D] "Design and Analysis of Auburn Dam,"

vol. 4, U.S. Department of the

Interior, Bureau of Reclamation, 1978.

[E] "Earthquake Ground Motions for Design

and Analysis of Dams," Von Thun, J. L. et al . , Earthquake Engineering and Soil

Dynamics II - Recent Advances in Ground

Motion Evaluation, Geotechnical Special

Publ ication

No. 20, American Society of Civil

Engineers, June 1988.

[F] "Design Criteria for Concrete Arch and

Gravity Dams," U.S. Department of

the Interior, Bureau of Reclamation,

Engineering Monograph 19, Denver,

Colorado, February 1974.

[G] Chopra. A. K.. and P. Chakrabarti, "The

Koyna Earthquake and the Damage

to Koyna Dam," Bulletin of the

Sei smol ogi cal Society of American,

vol . 63

,

No. 2, April 1973.

[H] Chopra, A. K., "Earthquake Resistant

Design of Concrete Gravity Dams,"

Journal of the Structural Division,

American Society of Civil Engineers,

June 1978.

190
Fig. 1 MAXIMUM DAM SECTION

Fig. 2 RESPONSE SPECTRUM

M 6.0 AT 2 KM - 5 % DAMPING

191
SYNTHETIC ACCELEROGRAM

M6.0 AT 2 KM - HORIZONTAL MOTION

3.76 i.OO «.2» TIME - SECONDS

7^0 «.7S

Fig. 3 SYNTHETIC ACCELEROGRAM

HORIZONTAL MOTION

Fig. 4 SYNTHETIC ACCELEROGRAM

VERTICAL MOTION

192
UPPER STILLWATER USING EFIOHI- STATIC * USSR S.O 8 2 KM

FROJCCTCD PUHN VID4 n?C»1 eLCMOTT 1 ID ELDIENT 212

Fig. 5 MODEL GRID

MAXIMUM DAM SECTION

UPPER STILLWATER USING EFIOHI- STATIC * USSR 6.0 6 2 KM

PTOXCTCO pum vicu rRcn axno^ i ro elemkt i67

Fig. 6 MODEL GRID

TYPICAL SECTION

193
eiSOr

MAXIMUM PRINCIPAL STRE5S-PSI

Fig. 7 PRINCIPAL STRESSES • MAXIMUM SECTION

WATER + DEAD LOAD + MCE

8130

7950

1 1 . I I I I I 0 100 200 300 400 »00 MAXIMUM PRINCIPAL STRESS-PSI

Fig. 8 PRINCIPAL STRESSES - TVPICAL SECTION

WATER + DEAD LOAD + MCE

194
ELEVATION 8095

400 r

z

00

- 100

—o WSE 8172, Time = 4.32 —^ WSE 8028.4. Time ^3.05

ELEVATION 8040

Z

CO

WSE 8172, Time =-5.19

WSE 8028.4, Time ^2.23

Fig. 9 VERTICAL STRESSES AT MAXIMUM SECTION

WATER + DEAD LOAD + MCE

195
Fig. 10 VERTICAL STRESS HISTORY - MAXIMUM SECTION

UPSTREAM FACE - ELEVATION 8040 - HIGH WATER

UPPER STILLWATER DAM - MAXIMUM SECTION

Stress History of Element 132 (Elevation 8040)

For USE 8028.4 + USSR M6.0 at 2 km, 10 percent damping, and 4 modes

irv i i

fx

O.OO US Z-SO 3.7& S.OO •.2ft 7.90 8.7S 10 00 TIME - SECONDS

Fig. 11 VERTICAL STRESS HISTORY - MAXIMUM SECTION

UPSTREAM FACE - ELEVATION 8040 - LOW WATER

196
UPPER STILLMATER DAM - MAXIMUM SECTION

Stress History of Element 167 (Elevation 8095)

For WSE 8172.0 + USBR M6.0 at 2 km, 10 percent damping, and 4 modes

J. 79 5.00 ».2S TIME - SECONDS

Fig. 12 VERTICAL STRESS HISTORY - MAXIMUM SECTION

DOWNSTREAM FACE - ELEVATION 809S HIGH WATER

UPPER STILLWATER DAM - MAXIMUM SECTION

Stress History of Element 167 (Elevation 8095)

For WSE 8028.4 +USBR M6.0 at 2 km, 10 percent damping, and 4 modes

N-/ iHH V

i 1\f

rv

O.OO L?» 2. 50 79 5 TIME -00 6. SECONDS

Z3 7. 50 6. 75 I0,(

I/) Ul , n;

.

»— > ID • <

>

Fig. 13 VERTICAL STRESS HISTORY - MAXIMUM SECTION

DOWNSTREAM FACE - ELEVATION 8095 - LOW WAl ER

197
Damage of Soil Liquefaction Caused by Earthquakes and

Use of Geomorphological Maps

by

Takekazu AKAGIRI*

(abstract)

Damage of soil liquefaction caused by big

earthquakes have been increasing recent years. The fundamental relationships between soil liquefaction and geomorphological units have been clarified. The author researched soil liquefaction caused by Niigata, Nihonkai -chubu

and Chibaken toho-oki earthquakes to review the relationships based on literatures and field

survey.And he concluded that (l)soil liquefaction

phenomena were the most remarkably caused in former river courses filled about several ten years ago and banked areas, (2)the qualitative

susceptibility to liquefaction for each geomor- phological unit has been interpreted using

geomorphological maps by Geographical Survey In- stitute (GSI) and (3)micro-geomorphoiogical maps are available for prediction of liquefaction.

[keywords] so i I liquefaction, geomorphological

maps, (licro-zonation, sub-surface and landform

1 . Introduction Big earthquakes have caused sandy soil liquefaction, collapses of slopes, and other

types of ground failures. Many cases of soil liquefaction have been found in the past. Traces

of liquefaction caused in 12c. to 19c.. was

found in old mounds in the ruins of 5 century in the subsurface layers, which sand boil had not reach the ground surface (Sangawa: 1987). And Kanto earthquake (1923, M=7.9) caused soil liquefaction extensively in the lower drainage

basin of the Nakagawa river (Gelological Survey

1925, Wakamatsu 1983). Also remarkable soil liquefaction was caused by Niigata earthquake

(1964:M=7.5), Miyagiken-oki earthquake (1978: M=7.4), Ni honka i -chubu earthquake (M=7.7),

Chibaken-toho-oki earthquake (1987 M=7.5)

(fig.K index map) and others. Most of plains in Japan have been formed by fluvial process. Material has been transported from mountainous slopes to the lower drainage

basin and deposited. Landforms and subsurface

ground have been formed by the same process.

Thus, it is possible to know subsurface ground

condition using micro- landforms.

In this paper, the author reports that it is easy to interpret and predict qualitative sus- ceptibility to the liquefaction using geomorphological maps based on the relationships be- tween landforms and soil liquefaction.

2. Soil Liquefaction Caused by Earthquakes

2-1. Niigata earthquake (1964. M-7.5) This earthquake induced various kinds of damages in Niigata and its surroundings: tunami,

Geographical Survey Institute

collapses and tilting of houses and buildings,

inundation, fire, sudden subsidence and rising

of ground surface, crack and sand boil with mud crater and etc.. Among them, soil liquefaction was caused in the extensive area in the lower drainage basin of the river Shinano.

GSI researched the damage using aerial

photographs and in the field and compiled maps showing aspects of damage caused by Niigata

earthquake and applied geomorphological maps at

a scale of i:iO,000 (1964: fig. 2, 3). According to the maps, liquefaction was

caused remarkably in numbers in the former river courses filled after 1926 along the Shinano

river, former river courses in the coastal

lowland between the Agano and Shinano, (The area

had been a river course during the past 400

years: Wakamatsu 1983), Niigata airport located at artificially changed dunes, interlevee

lowlands, edges of natural levees with high

ground water level and banked roads in the

lowlands (tab. 1 ,f ig.3).

Subsurface ground along the Shinano was

displaced laterally toward the river about 8m

(one side)in the both sides of the river by soil liquefaction and the width of the river became narrowerdlamada et al. 1986). This is clearly

found using a set of aerial photographs by GSI, taken before and after the earthquake.

In the subsurface of the damaged areas. Following characteristics were common to the areas in the subsurface of damaged areas:

(l)sand deposited thick, which average diameter

of the grain size is 0.1'-0.5mm, (2)relative

density is low and (3)ground water level is high. But, there was no damaged area by soil liquefaction in well developed dunes and

natural levees, where ground water level is low and soil is compacted (Japan association of ar- chitecture 1964, Aoki, Ogawa, Vagi 1974, GSi:

1964).

2-2. Nihonkai -chubu earthquake (Middle Japan Sea eq.M=7.7)

This earthquake caused intensive damage by

tunami from Aomori to Akita: maximum 15m high

at the coast, several ten km from north to south along the coast, inundated land area 1km deep

inland, various types of ground failures of ground, destruction of houses and buildings,

destruction of lifelines and others. Liquefaction phenomena were caused in many places in the lowlands. Long embankments sub- sided in the surrounding of the lake

Hachi rougata. In many places, liquefaction was observed in similar landform units to Niigata in former river courses of the river Omono, banked

198
area, edges of sandy landform units (natural

levees, sand dune). Lateral displacement of paddy fields was observed(f ig.-4, 1984). And at the edges of sandy landforms, sand

was lost at the base of houses too much and

damages of houses became larger. But only one house was not damaged at all. According to author's Interview with the owner, the house was built up on the basis of enough piles (deep and many) into marshy land because the owner had

known that subsurface ground condition had been

the worst for housing.

2-3. Chibaken-Toho-Oki Earthquake (1987,(1=6.7) This earthquake hit the neighbor area of Tokyo and caused intensive damages in Chiba and

Ibaraki prefectures. Damages are destruction of houses more than about several ten thousands,

destruction of life lines, damages of roads,

liquefaction in alluvial lowlands and fill along

Tokyo Bay, former river courses and banked area

in the lowlands of Kujukuri plain, former river courses in the lower drainage basin of the Tone river, and the reclaimed area by draining the

lake in Itako Town. And slope collapses were in- duced in the hill slopes of the Kujukuri coastal

area. Liquefaction phenomena were caused in many places. Fig. 5, 6, 7 are examples in the lower drainage basin of the Tone river. Comparing

with liquefied sites and the geomorphological map drawn at a scale of 1:25,000 by GSI,

liquefaction was caused in the alluvial

lowland(one area), a combination with former

river course and slip off slope (5 areas). Most

of sand boil were formed in the former river

courses. No soil liquefaction phenomena was not observed in undercut slope. The author researched histories of these liquefied sites using the whole editions of former topographic maps since 1880. According to the maps, liquefied sites had been river courses

several ten years ago. And after the construction of short cut in 1915, these sites remained

as oxbow lakes where water chestnuts grew thick. And these were filled using river sand of the new channel of Tone river with sandpumps from

1960 to 1967. Later, partly river sand was sold out and backfilled using hilly sand. Water-chestnuts in the area had deposited

at the river floor before fill and came out with sand boil this time. Crack type of sand boil continue intermittently and side by side about

several hundreds meters long (fig. 5). The dis- tribution coincided with the backfilled hilly

sand area. According to the boring data, which was reseached after the earthquake, by the science

and engineering research laboratory of Waseda univ., surface geology consists of the

followings: 0-6m:hill sand, river floor sand,

6-10ra:fine sand, 10m-:very fine sand. And sand

boil consists of fine sand and very fine sand.

Then, sand boil came out from deeper layers

below hilly sand layer. Most of liquefaction were caused in the former river courses. Grain sizes are not well sorted. Sand boil was not observed In the former

river courses which were filled up by natural

processes or by gradual fill without sandpump.

3. Subsurface Condition and Geomorphological Maps

3-1. Geomorphological mapping and subsurface con- dition

Soil liquefaction is caused by the combina- tion with (l)seismic intensity and duration,

(2)density of sand (void ratio, relative den- sity, H value), (3)grain size distribution,

(4)ground water level, (5)depth of sand layer.

And the phenomena change depend on the combination with these. Among them, except (1), others relates to subsurface condition and it shows

that subsurface is indispensable to predict soil

I iquefaction si tes. Most of plains have been formed by fluvial process in Japan. Landform complex in the upper drainage basin has not changed in a period that microlandforms have been formed in the lower plain in usual. For the reason, in the lower

plain, same material, which is transported from

the upper drainage basin, has been deposited in specified sites repeatedly and microlandforms

have been formed. This process decides material and types of subsurface ground condition whether a fan develops to a big one or a little scale of boulder fan, and whether the fan consists of gravels or fine sand or silt. Similar type of subsurface have similar characteristics of sub- surface condition . Vakamatsu et. al (1980)

proposed that qualitative susceptibility to soil liquefaction was decided by the combination with types of plain, characteristics and scale of landform units.

As above mentioned, microlandforms show not only details of relief of plain, but also

geomorphological genetic history and have indis- pensable relations to subsurface condition which are surface geology, soil condition, ground

water and so on. Thus, it is naturally possible

to make clear subsurface conditions using

microlandforms.

As an example showing the relations, there

is the distribution of cracks with sand boil in the lowland of Nakagawa river drainage basin in Saitama prefecture to the north of Tokyo, which

was caused by 1923 Kanto earthquake (M7.9) in (Gelogical .Survey: 1925). The distribution coin- cided with the distribution of natural levees

along former river courses. And it is clear that

the numbers of liquefaction varied in each kind

of geomorphological unit. And recent years,

damages of soil liquefaction were caused in ar- tificially changed landforms and subsurface.

199
Oya, Wakanatsu et al.(1982) compiled a geomorpho logical map for prediction of flooding

and liquefaction in Shonai plain and made clear qualitative susceptibility to liquefaction at each geomorpho logical unit based on the research

how frequently liquefaction was caused at each geomorpho logical unit subject to seismic ground motion of the J.M.A. intensity V (J.M.A.V cor- responds to Modified Mercalli intensity \i

approximately) using the past earthquake. Three months after compilation of the map, Nihonkai- chubu earthquake caused liquefaction near the spring at dune edge where the map had predicted. Kotoda et al. (1988) clarified peak ground

velocity of the past 19 earthquakes at liquefied and non- I iquef ied sites on various geomorpho logical units to make clear the quantitative

susceptibility to liquefaction. And they got

the result of fig. -8, 9 which showed critical

values causing liquefaction. This does not con- flict with the result by Oya et al..

3-2. Prediction of sites using geomorphological

maps Above mentioned geomorphological map was compiled based on the method of Oya's geomor- phological map for prediction of flooding at a scale of 1:25,000. And geomorphological map series (1:25,000 land condition nap series) by

GSI was also based on his method with additional conceptions and legend for prediction of flooding at the beginning. But at this moment it has been being prepared for various purposes including ground failures by earthquake.

A geomorphological map is a map showing

microlandforms based on aerial photo interpreta- tion and field check to classify morphological

characteristics. Ceomorphologicaily, the objective area is classified into homogeneous units

by the combination with form, material, genetic

process and period the landform was formed. Classification consists of three classes:

Large units are mountainous area, hi 1 1 -and-table land, and lowland. Middle units are valley

floor, coastal and deltaic plain in the case of lowland. And detailed units are a former river course and a natural levee in the case of deltaic lowland.

In technology, micro-zoning maps for pre- diction of liquefaction decide susceptibility to liquefaction at each 500m mesh using boring data (geology and N value). The method is a typical

quantitative method with an enormous numbers of boring data and the researcher must analyse data

at each mesh. And the reliability becomes lower

in an area with few boring data. The method is not a preliminary method, but a fundamental re- search.The method based on geomorphological map- ping is qualitatively. And (l)it is possible to classify pretty extensive area based on aerial photointerpretation in a short period at low

cost. And many geomorphological maps covering most of plains have been prepared at middle

scales. (2)in comparison with a mesh map, it is possible to draw more suitable boundaries which

corresponds with actual boundaries of subsurface

conditions using geomorphological maps. (3)even

if users do not have enough knowledge of soil engineering and geomorphology, it is easy to in- terpret subsurface conditions and susceptibility

to liquefaction. (4)it is possible to research

general condition at a scale of 1:50,000 first, and if necessary, more detailed work can be done

at a scale of I : 10,000 or 1:5,000. And quantitative work can be done after it. It means that

this method has characteristics as a preliminary research. The National Land Agency has been

preparing micro-zonation maps for prediction of liquefaction with geomorphological mapping.

4 -Cone I us ions (1) The author researched a few earthquakes

which caused soil liquefaction and reviewed the relationships between geomorphological units and soil liquefaction. Based on the research of Chibaken-toho-oki earthquake, it was clarified

that soil liquefaction was caused in deeper

layer than 4m and in the former river courses

and fill about several ten years ago. And these sites were shown as specified landform units in the geomorphological maps published by GSI. (2) The susceptibility to liquefaction for each

geomorphological unit based on geomorphological

mapping does not conflict with the result that Wakamatsu et.al clarified qualitative susceptibility to liquefaction at every geomorphological unit based on peak ground velocity. And it

is reasonably possible to explain conceptions

based on fluvial process. Existing geomorphological maps are avail- able to m i crozona t i on for prediction of

liquefaction. And the microzonation based on geomorphological mapping is effective as a prel iminary research.

Acknowledgment The author would like to thank Mrs. Kazue

Wakamatsu, Waseda Univ. for her helpful suggestion.

References (E): in Engl ish Akagi ri ,T( 1988): Damages Caused by Chibaken-toho

Earthquake. Report of GSI(Jihou) vol.67 Oya.M ed.(1983): Method and Development of Geomorphological Mapping. Kokon-shoin Tokyo

Oya, Kotoda, Wakamatsu, Kubo(1982): A geomorphological map for prediction of flooding and liquefaction in the Shonai plain.

GSI (1976): Land condition map series and document

of Sawara

GSI(1964): Actual conditions of damages caused

by Niigata earthquake and land condition with document i: 10,000 2sheets

200
Kotoda, Wakamatsu( 1988): Liquefaction phenomena

and resulting damage to structures during the damages caused by Chibal<en-toho Earthquake. Soil and Foundation vol 36-12, ppl9-24

Kotoda, Wakamatsu and Mase(1988): Liquefaction

phenomena caused by 1987 Chibaken-toho

earthquake, part 2. 23rd meeting of soil en- gineering proceedings. pp959-962

Sunaga, Kunaki (1982) :Subsurface damages caused

by earthquakes and geomorpho logical mapping.

Report of GSI(Jihou) No. 56 pp24-31 First Geographic Division, GSI(1984): Geomorpho logical map and Damages Caused by Nihonkai chubu earthquake

Geological SurveyC 1925): Report on Kanto earthquake, part 1. 240p. Nakano and Oya(1960):The relationships between

landforms and high tide flooding, sokuryou. 1960 Hamada, Yasuda, Isoyama and Emoto(1986): Per- manent displacement of ground caused by

liquefaction. Study on liquefaction induced per- manent ground displacements. Civil Association

for the Development of Earthquake Prediction

PP87

Sangawa( 1987): Traces of earthquake found in Hotarudani ruins in Otu city in Shiga pref. Chishitu News No. 390. Wakamatsu,K( 1983): Damages caused by earthquakes

and geomorpho logical maps. Oya ed. "Method and development of Geomorpho logical mapping" pl57-

169 Wakamatsu , K( 1 980) : Geomorpho I og i ca I cons i de ra t i on on site conditions of soil liquefaction caused

by earthquake. Asian Profile vol.8, No. 3, 279- 297 (E) Watanabe, Iwata, Yasuda, Sato and Ohashi (1983): I8th meeting of soil engineering, proceedings. PP567-568

Chiba

f ig.-l

1 mm

201

203
i^'^m(Dm^Wi G.L. before eaarthquake

m^m(Dm^m G.L. after eearthquake

-^^^ m m ti cracks

^'P . sand boil

ground crack with vertical gap

c=^ ±m,(D^$myD[^ lateral displacement

boundary of paddy field displaced by the eq,

+ 5.4-75.0

fig. 4- Landform change (lateral displacement) at

Gomyoukou near Lake Hachirougata in Akita pref.

Caused by Nihonkai-Chubu Earthquake, 1983"

by First Geographic Division of GSI(1983)

204
fig. -5 Distribution of Sand Boils in the Lower

Drainage Basin of The Tone River

ill

"ii

205
fig. 6- Distribution of Sand Boils in the Lower

Drainage of the Tone River

A

i?IIM«e::*:nii^2.85mx2.1m, m%

206
207
ALLUVIAL FAN

NATURAL LEVEE

RIVER CHANNEL

BACK MARSH VALLEY PLAIN

DELTA

SAND DUNE, LOWLAND BETWEEN

SAND DUNES

SAND BAR

RECLAIMED OR DRAINED LAND

I I I

I

1 r—I—1—I J I J o coriDaiX)

OO (EDCOCD

+++-H-H+++ -H*-+H+)- +

NO LIQUEFACTION I \ (TOSD CtmD

+ -H-+ + 4-Ht- ++ + \ LIQUEFACTION

O CIDO OO

O OO + ++• + + +

GDcrx5<iDrca3H3 COX)

1 1 1 1 I I I 10 20 50 100

- PGV kine

(cm/sj

Fig. 8 Peak ground velocity at liquefied

and non-liquefied sites on various

geomorphological units

fig. 9- Peak ground velocity which causes liquefaction

landf oris peak ground velocity

f i 1 1 i 1 1

1

reclaiaed land

by draining

natural levee

riupr 1 L V \Z L pniir<%p ^UULo^

edge of dune

interdune lowland

15cib/s

backmarsh

valley floor

delta surface

25ciii/s

beach ridge

fan 35cin/s

geoHorphology

existing naps literature

coiaparison with

similar plains

aerial photo

interpretation

f ieldcheck

interview of liquefaction

praparation of

prelisinary nap

collection of earthquake

damage

compilation of

geomorphological

map

distribution map

of liquefaction by

past earthquakes

clarifying the relationships

between landforins and

liquefaction

1

supplementary work of geomorphological map

decision of susceptibility

to liquefaction for each

geosorpholgical unit

completion of geomorphological map

for prediction of

1 iquefaction

fig. 10- A flowchart to compile a geomorphological map for prediction of

liquefaction

208
Dynamic Defonnation Characteristics of Compacted Rockfills

by Cyclic Torsional Simple Shear Tests

by

Norihisa MATSUMOTO Nario YASUDA Masahiko OHOKUBO ^ Yasushi KINOSHITA

ABSTRACT

Large-scale cyclic torsional simple

shear tests (CTSS) were performed to

investigate the dynamic deformation

characteristics of compacted rockfills

(gravels) at wide shear strain range.

The effects of confining pressure/ void

ratiO/ principal stress ratio and cycle

loading frequency on the shear

modulus and damping ratio are evaluated. It has been found that shear

modulus is expressed as the function of

void ratio and confining pressure for

compacted rockfill as proposed in the

past studies for sands/ and damping

ratio can be expressed as a linear

function of void ratio. The influence

of principal stress ratio to shear

modulus in the cyclic torsional simple

shear test(CTSS) was proven to be

different from that at cyclic triaxial

test(CTX). Furthermore/ the strain

dependent curves of normalized shear

modulus and damping ratio by CTSS for

gravels show similar inclination

comparing to those from CTX.

1 . INTROD U CTION

It is important for the rational and

economical design/ and construction of

rockfill dams to study the dynamic

properties of compacted rockfills

(gravels )/ which occupy a large part

of the dam body and contribute greatly

the stability. Furthermore it is desirable that these properties be

obtained under stress-strain state

similar to that of the field.

The cyclic triaxial test using solid

specimen of about 30 cm in diameter and

60 cm in height has been frequently

conducted due to easy handling.

However/ the triaxial test/ in which the

principal stress axes are fixed in

vertical and horizontal directions/

does not simulate necessarily the

stress-strain state of the dam body

during earthquakes.

During earthquakes/ the shearing

vibration is apt to be predominant in

the dynamic stress-strain state of

rockfill dams with a continuous

rotation of principal stress axes. Simple shear test apparatuses of SGI/

NGI and Cambridge types can reproduce

the stress-strain state of the field

elements subject to shear

deformations. However/ the effects of

the end surface of specimen/ namely/

the shortage of transmitted shear

stress and the stress concentration at the inside of specimen are conspicuous

in this apparatus. On the other hand / these effects are negligible in CTSS

owing to its unlimited cylindrical

shape.

A large number of hollow cylindrical

torsional simple shear tests have been

performed for static and dynamic

strength and deformation

characteristics of sands to date, only

a few tests have been carried out on

rockfills. In order to investigate the

dynamic deformation characteristics of

rockfills (maximum particle size of 38.1

mm) corresponding to grained

stress-strain state in the field/ the

research was conducted for a wide

shear strain range by using

large-scale cyclic torsional simple

shear device with hollow cylindrical

specimen.

2. LARQ E-SCAL E CYCLIC TORSIONAL

SIMPLE SHEAR DEVICE

All experiments were performed using

large-scale cyclic torsional simple

shear device. In this device a hollow

cylindrical specimen of 80 cm in height/ can apply independently to the

specimen axial stress/ confining

pressure and torsional shear stress.

Furthermore/ triaxial compression and

extension tests can also be carried out

by preparing the solid specimen of 40

cm in diameter and 80 cm long in this

devi ce. Fukushima/ et al. (1980) investigated

the ratio of the outer diameter to

height of hollow cylindrical specimens

using four different heights of the

1 . Chi ef /Fi I I dam Division/Public Works

Research Institute/the Ministry of

Construction/TSUKUBA City/Japan

2. Research Engineer/ ditto

3. Engi neer/ ditto

4. Consulting Engineer/ C.T.I.

Engineering/ Co. /Ltd. Formerly visiting

r esear cher of P. W.R. I

.

209
preparing the solid specimen of 40 cm

in diameter and 80 cm long in this

devi ce. Fukushima/ et al. (1980) investigated

the ratio of the outer diameter to

height of hollow cylindrical specimens

using four different heights of the

specimens/ with the same outer and

inner diameters for dry sands. They

reported that the height of specimen was desirable to be larger than the

outer diameter to eliminate the

disturbance of end constraints. Thus/

the ratio of 1.0 has been applied to

this large-scale torsional simple shear

device as the proportion of a height to

a diameter.

3. TEST CONDITIONS AND PROCEDURES

Quartz andesite was used for the

specimen materials in this study. The

materials were transported from the

quarry site of a rockfill dam now under

construction to the laboratory of

P. W.R.I. The grain size distribution of

a specimen is determined so as to

resemble that in field/ and illustrated

in Fig 1. The maximum grain particle

size is 38.1 mm resulting from the

requirement that maximum particle

diameter must not exceed 1/5 of the

thickness of the specimen diameter.

All exper iments wer e performed under

the air-dried and drained conditions.

Table 1 shows test cases conducted this

study.

To accommodate the 10 th hysteresis

loop of stress-strain under steady

state at every 0.01 second/ an analog-digital conversion system and

data storage system with memory

controlled by a personal computer are

used for data acquisition.

4. EXPERIMENTAL RESULTS AND ANALYSIS

4. 1 Dynamic Test Results

Shear strain dependent curves of

modulus G and damping ratio h with void

ratio ec=0.42 and stress ratio cr a a

= 1.0 are illustrated in Fig. 2. The

smooth curves of shear modulus G were

obtained at the strain range of 2XlO~®

^SXlO"'*. Usually/ dynamic deformation

characteristics at the shear strain

levels less than 1 X 10"^ was obtained

by resonant column test/ and those at more than 1 X lO"'^ was from cyclic

triaxial test (CTX). Both test methods

rarely coincided with each other. In this study/ dynamic properties at wide

shear strain range including strain

less than 1 X 10~^ are continuously

gained by CTX due to the usage of the

ultra-sensitive displacement

transducer (Gap sensor).

The effect of confining pressure cr on damping ratio were found in the case of sands/ however it has not been

obvious in rockfills. Otherwise the

effect of cr c on damping ratio was recognized at cyclic torsional simple

shear test (CTSS) of this study as described later.

Fig. 3 indicates the relation between

shear modulus G and void ratio after a consolidation ec. G is divided by the

function of confining pressure ( cr c)'" in order to eliminate the effect of 0" ^ itself. It is distinct that the void

ratio eo increases/ the shear modulus

G decreases. The relation of G to Cc

for rockfills is well expressed in Eq. 1 proposed by Hardin and Richart (1963)

for shear modulus of sands at minute

shear st r ai n.

(2. 17-eo)2

Where/ K = constant depending on shear strain/ m = power related to

confining pressure (depending on shear

strain)/ cr = effective confining

pressure.

CTSS results of this study are plotted around approximate straight

lines from regression analysis as shown

in Fig. 3. Kokusho/ et al. (1980) also

proposed the same slope of approximate

straight lines for crushed and round

rocks by CTX/ so it can be confirmed

that the relation between shear

modulus G and void ratio e© for

rockfills is independent of testing

methods/ and can be explained as Eq. 1. Fig. 4 indicates the relation of shear

modulus G to confining pressure cr c both in logarithmic scales with shear

strains as parameters. G at ordinate is divided by Eq. 1 to remove the effect

of void ratio. G decreases with the

increase of shear strain and increasing

with the rise of CT c. The incline of

linear regression line drawn in Fig. 4

becomes gentler according to the

smaller shear strain range as well as

r esu Its f r om CTX.

Fig. 5 shows the power/ m, relating

shear modulus G to confining pressure.

The power m concerning confining

pressure cr ^ i s a function of shear

strain. CTX results of gravel as are

210
plotted together in this figure with

test data of the sand from Kokusho/ et

al. (1980) and Iwasaki, et al. (1980). It can be seen that power m of rockfills

is distributed from 0.65 to 0.85

irrespective of testing method/ and is larger than that of sands at wider

shear strain range. Consequently/

shear modulus 6 of rockfills can be

displayed by Eq. 1 at not only very

small but also large shear.

Fig. 6 shows the relation of damping

ratio h to void ratio ec. Damping ratios

are also normalized by the function of

the confining pressure ( 0" c)" to

eliminate the effect of the confining

pressure in the same way as the case of

shear modulus. The relationship has a precise inclination to increase

according to the rise of void ratio at shear strain range of 5X 10"^ 5X

10~^. This tendency is attributed to

the increase of energy loss by small

plastic deformations of the specimen in

proportion to the higher void ratio

against the same cyclic load. The

relationship between damping ratio and

void ratio can be expressed by the

f 0 I I owi ng equati on.

h = L( r ) /S ( cr^)-

/9 = 0.4 + 1,5 ec Where/ L = constant depending on

shear strain/ n = power related to

confining pressure (depending on shear

str ai n).

Fig. 7 illustrates the relation of

damping ratio \\ / ^ to confining

pressure 0" o. Damping ratio is normalized by the function of void

ratio ec to eliminate the influence of

Be as the case of shear modulus. The

slopes of linear regression lines

become steeper accompanying with the

smaller shear strain in opposition to

the relation between G and C c. The

above-mentioned inclination of damping

ratio is not often apparent in CTX

testi ng.

Principal stress ratio ( CTi/CTa, in

which cr , = major principal stress/ CJ a

= minor principal stress) of a rock

zone at rockfill dam body ranges from

1.0 to 2.0 according to monitoring of

observational instruments at the dam

body/ and becomes larger inside the

body. In this study/ CTSS is also

performed under two different

anisotropic consolidations of cr i/ 0" 3

= 1.5 and 2.0 besides isotropic

consolidation. The relation between

principal stress ratio 0" i/CTa and Rs/

which is the ratio of shear modulus G

against shear modulus at cr i/CTs of 1.0

both under the same mean principal

stress conditions/ is indicated in Fig.

8. Figure is arranged for shear strain

of 2X 10~^ and four levels of mean

principal stress cr ^ - 1.0/ 2.0/ 3.0 and

4.0 kgf/cm^. Rs is less than 1.0 about

cyclic torsional simple shear test

(CTSS). The results of cyclic triaxial

test (CTX) by this apparatus are also

shown in Fig. 8/ and Rs ranges from 1.2. to 1.4 illustrating the reverse

inclination of principal stress ratio in

comparison with CTSS. Test result of

CTX with solid specimen of 0 30 cm in

diameter and 60 cm high is conducted by

Matsumoto/ et al. (1986) using the same quartz andesite and plotted in this

figure.

The influence of 0" i/aa to damping

ratio is not so clear as that to shear

modulus. It is estimated that there is not a significant effect of cr i/CTa upon

the dam ping.

It is generally known that the clay

materials are not affected by cyclic

frequency of dynamic load except for a large deformation as a creep. Hara and

Kiyota (1979) concluded that shear

modulus of clean sands is constant

value at the frequency range of 0.1 ^

10.0 Hz. In this study the dependency

of frequency for rockfills is investigated about the lower range of

0.01 ^ 0.2 Hz by reason of the limited

capacity of testing device. Tests are accomplished for four levels of

confining pressure and strain range of

2XIO-® ~ SXlO-'^, and it is recognized

that almost similar shear modulus are observed at each pressure level, (see

Fig. 9)

Hardin and Drnevich (1972) proposed

that the relation between shear

modulus G and damping ratio h is independent of void ratio ec/ confining

pressure cr c and shear strain/ and

simply expressed as Eq. 3.

G

h = hmax ( 1 - ) (3)

Go

Where/ Go = maximum shear modulus/

hmax = damping ratio at G=0. Fig. 10

shows the relation of normalized shear

modulus G/Go to damping ratio h for

this testing. Results for specimens at three different void ratios are plotted

with their linear regression lines/ and

the line from the average of whole

data. A least squares regression fit is performed to obtain these linear

211
regression lines. Data of each void

ratio scatters around the linear line

independent of shear strain and

confining pressure 0" c/ and high

correlation between data and lines are

evident. However/ hmax which shows

the slope of linear line/ becomes

larger according to the increase of e©

as is corresponding to the relation

between them in Fig. 6. It has also been

shown that hmax of sandy soils are determined irrespective of e© as for

average line of several kinds of

natural sands drawn in this figure

(Tatsuoka/ et al. (1978)). Though/

gravels have a different inclination

according to ec and indicate lower

values comparing with the average line

of sands. The effect of principal

stress ratio were not evident in this

study.

4.2 $trai n Dependent Curves of

Rockf i I Is The average degradation curve of

normalized shear modulus observed in

this laboratory test is shown in Fig. 11 together with the results of rockfills

from other researchers. Matsumoto/ et

al. (1985) conducted cyclic triaxial

test (CTX) using crushed angular

gravels with the specimen of 30 cm in

height and 60 cm in diameter.

Reconstituted samples of Tokyo gravel

of 90 mm maximum particle size were

tested with the same size of specimens

in CTX by Hatanaka, et al. (1988). The

normalized average degradation curve

after Seed/ et al. (1984) consists of

results from angular and rounded

rockfills of several earth and rockfill

dams. Hynes/ et al. (1988) conducted

tests using rounded rockfills from

Folsom Dam. The curves obtained by

Japanese researchers are about 30

percent higher than those from Seed/

et al. (1984) and Hynes/ et al. (1988) at strai n of 1 X lO""^. The strain dependent curves of

damping ratio indicate similar

inclination in shear strain range of lX

10-^ to 5X10"^/ though CTX results are expected to show slightly smaller

values in larger shear strain except

for the results from Seed/ et al.

(1984).

These divergences may come from the

differences of materials/ and test

equipment and procedures used in both

countries/ so the further study seems

to be necessary in future.

5. CONCLUSION

Air-dried rockfills were tested in a large-scale hollow cylindrical cyclic

torsional simple shear apparatus in

drained condition. The experiments

were conducted under stress

controlled condition and about dynamic

characteristics at wide shear strain

range including less than 1 X 10~^. the

following conclusion can be drawn for

rockf i Ms in this study:

1. K • (2. 17-ec)2/(l + eo) • ( cr ^)m can express shear modulus at wide shear

strain including initial shear modulus.

Power m is distributed from 0.65 to 0.85

independent of shear test method and

higher than that of sand.

2. Though the shear modulus of cyclic

torsional simple shear test (CTSS) tend

to become larger with the increase of

principal stress ratio ranging from 1.0

to 2.0/ ones of cyclic triaxial

compression test (CTX) have an inclination to be smaller under the same stress ratio and mean principal

st r ess.

3. The relation between damping ratio

and void ratio is obvious at CTSS and

explained as the linear function.

4. Shear modulus relates to damping

ratio lineally as proposed by Hardin

and Drnevich (1972) and linear

regression line has a gentle slope

compar i ng to sands.

5. The strain dependent curves of

shear modulus obtained in Japan is 30 %

higher than those in the United States.

References

1. Fukushima/ S./ Nakajima/ K. and

Tatsuoka, F (1980). "Drained Shear

Characteristics of Sand in Torsional

Simple Shear Test"/ The proceeding of

16th Annual Meeting/ JSSMFE/ pp.

102-105 (in Japanese)

2. Mara/ A. and kiyota, Y./ (1979).

"Study on Dynamic Properties of Soils

for Dynamic Analysis of Foundation"/

The proceeding of 14th Annual Meeting/

JSSMFE ( in Japanese)

3. Hardin/ B. 0. and Drnevich/ V. P./

(1972). "Shear Modulus and Damping in

Soils/ Design Equation and Curves"/

Journal of Soil Mechanics and

Foundation Engineering, ASCE, Vol, 98/

No. SM7, pp. 667-692.

4. Hardin/ B. 0. and Richart, F. E. Jr.,

(1963)/ "Elastic Wave Velocities in

Granular Soils"/ Journal of Soil

Mechanics and Foundation Engineering/

Proceeding of ASCE/ Vol.89/ No. SML

212
pp. 33-35

5. Hatanaka/ M., (1975/1976)/ "Dynamic

Behavior of Sand Obtained in Ring

Torsion Shear Test"/ Reports of

Committee on Dynamic Properties of

Soils/ Building Research Institute/

Ministry of Construction (in Japanese)

6. Hatanaka/ M./ Suzuki/ Y./ Kawasaki/

T. and Endo/ M./ (1988)/ "Cyclic

Undrained shear Properties of High

Quality Undisturbed Gravel"/ Soil and

Foundation, Vol.28/ No. 4/ JSSMFE,

PP57-68

7. Hynes/ M. WahL R. E./ Donaghe/ R

T. and Tsuchida/ T./ (1988)/ "Seismic

Stability Evaluation of Folsom Dam and

Reservoir Project/ Report 4/ Mormon

Island Auxiliary Dam - Phase 1"/ Technical report GL-87-14/ Department

of the Army/ WES/ Corps of engineers

8. Iwsaki/ T./ Tatsuoka/ F. and Takagi/

Y./ (1976)/ "Dynamic Properties of Sand

at Wide Shear Strain Range"/ Technical

Memorandum No. 1080/ Public Works

Research Institute/ Ministry of

Construction/ Japan (in Japanese)

9. Iwsaki/ T./ Tatsuoka/ F. and Takagi/

Y./ (1980)/ "Experimental Study on Dynamic Deformation Properties of the

Ground (II)"/ Technical Report No. 153/

Public Works Research Institute/

Ministry of Construction/ Japan (in

Japanese)

10. Kokusho/ T. Sakurai/ A. and Esashi/

Y./(1979)/ "Cyclic Triaxial Test of

Dynamic Soil Properties for Wide Strain

Range"/ Central Research Institute of

Electric Power Industry/ Civil

Engineering Laboratory/ No. 379002/ (in

Japanese)

11. Kokusho/ T. Esashi/ Y. and Sakurai/

A./ (1980). "Dynamic Properties of

Deformation and Damping of Coarse

Soils for Wide Strain Range"/ Central

Research Institute of Electric Power

Industry/ Civil engineering

Laboratory/ Report No. 380002/ (in

Japanese)

12. Matsumoto/ N. Yasuda/ N. and

Ohkubo/ M. (1986)/ "Measurement of

Dynamic Poisson's Ratio of Toyoura

Standard Sand"/ Proceedings of the

21st Annual Meeting/ JSSMFE/

pp. 555-556. (in Japanese)

13. MatsumotO/ N. Yasuda/ N. and Shiga/

M./ (1985)/ "Behaviours of a Rockf i I I Dam during Earthquakes"/ The 17th

Joint Meeting of U.S. -Japan Panel on

Wind and Seismic Effects/ UJNR

14. Matsumoto/ N. Yasuda/ N. and

Ohkubo/ M.. (1986)/ "Dynamic Shear

Modulus/ Damping RatiO/ and Poisson's

Ratio of Coar se-Gr ai ned Granular

Materials"/ Civil Engineering Journal/

Vol. 28/ No. 7/ Public Works Research

Institute/ Ministry of Construction/

Japan/ ( i n Japanese)

15. Matsumoto/ N. Watanabe/ K. and

Ohno/ K. (1984)/ "Dynamic Property

Tests for Sands and Rockfill Materials

by using Large-Scale Shearing

Apparatus - Hollow Cylindrical Simple

Shear and Cyclic Triaxial shear Test"/

Technical Memorandum No. 2132/ Public

Works Research Institute/ Ministry of

Construction/ Japan (in Japanese)

16. Ohoka, H. and Itoh/ K./ (1979)/

"Stress-Strain Behavior of Dry Sand

and Normally Consolidated Clay by

Inter-Laboratory Cooperative Cyclic

Shear Tests"/ The 11th Joint Meeting of

U.S. -Japan Panel on Wind and Seismic

Effects/ UJNR/ pp. 424-453

17. Seed/ H. B./ Wong/ R. T. Idriss/ I. M. and Tokimatsu/ K./ (1984)/ "Modulus and

Damping Factors for Dynamic Analysis

of Cohesionless Soils"/ Report No. EERC

84-14. Earthquake Engineering Research

Center/ University of California/

Berkeley/ Ca.

18. Suzuki/ Y./ Sugimoto/ M./ Babasaki/

R. and Kakita/ E. (1976)/ "Measurement

of Poisson's Ratio by Means of Dynamic

Triaxial Apparatus"/ Proceedings of the

11th Annual Meeting/ JSSMFE/ pp. 415-418

(i n Japanese)

19. Tatsuoka/ F., Iwasaki/ T. and

Takagi/ Y. (1979)/ "Hysteretic Damping

of Sands and Its Relation to Shear

Modulus/ Soil and Foundations/ Vol. 18/

No. 2/ June/ 1978

213
Table 1. Test Cases Conducted (CTSS and CTX)

Initial

Void

Ratio

Relative

Densi ty

(%)

Principal

Stress

Ratio

Confining

Pressure

( kgf/cm^)

Cycl ic

Frequency

(Hz)

1 0.42 70.9 1.0

2 0.47 59.2 0.2

3 0.52 47.4 1.0 2.0

4 0.42 70.9

3.0

0.2, 0.1

0.05, 0.01

5 0.42 70.9 1.5

0.2

6 0.42 70.9 2.0 4.0

Particle Size (mm)

Fig. 1 Grain Size Distribution Curve of Specimen

214
5000

UOOO

3000

2000

1000

o -

cI I I I I I I I \ cri/(T3 o 1.00 1.00

2.00 1.00

A 3.00 1.00

o ^.00 1.00

Shear Strain J"

Fig. 2(a) Strain Dependent Curves of Shear Modulus G (CTSS)

10 T—rrT"

Cm cXi /era 1.00 1.00 2.00 1.00 A 3.00 1.00 O ^.00 1.00

o

-B- 10 -6 10-5 2

Shear Strain

5 10 -4

r

5 10

Fig. 2(b) Strain Dependent Curves of Damping Ratio h (CTSS)

215
2000

0.35 0.^0 0X5 0.50 0.55

Void Ratio after Consolidation ec

Fig. 3 G/OTc'" versus Void Ratio ec (CTSS)

2000

^ 1000

500 -

200-

100

0.5 1 2 5 10

Confining Pressure^c (kgf/cm )

Fig. 4 G/a ver sus Confi ni ng Pressure 0" « (CTSS)

216
1.0

0.5

Matsumoto, et al . (1984)

CTX f630 X H60

'Iwasaki , et al . (1980)

Sand

Kokusho, et al . (1980)

Sand

O CTSS, «i80-40 X H80

A CTX 040 X H80

10 10 10"

Shear Strain If

10 -3

10

Fig. 5 Power Number m versus Shear Strain r (CTSS and CTX)

035 0^0 0.^5 0.50 0.55

Void Ratio after Consolidation ec

Fig. 6 h/OTo" versus Void Ratio ec (CTSS)

217
QJ

10 —Q—0—r=5xio

+ —g— 2x10 ^ 5

- c5

II

.—

.

2 - o ec=0. 42 - A ec=0. 47 - ec=0. 52

1

1 I

1 1 1 1

1 I I

I I I 0.5 1 2 5 10

Confining Pressure c (kgf/cm )

Fig. 7 h//9 versus Confining Pressure cr c (CTSS)

1.4

O

1.2

1.0

0.8

<1> T

as

(Tm \>

\

1.0

2.0

3.0 m

4.0 o <I> 6.0

/

'/

/

/

/ y

CTSS, 080-40 X H80

CTX 040 X H80

flatsumoto, et al CTX 030 X 1160

7

( 1986)^

1.0

^ -O1.5 2.0 2.5

Principal Stress Ratio

^^^Z

Fig. 8 Rs versus Principal Stress Ratio <j J u s, (CTSS and CTX)

218
4000

- 2000

o

o

0.01

-o

r =ix 10-"

r =5X iQ-"

0.05 0.10

Cycl ic Frequency (Hz)

0.20

Fig. 9 Shear Modulus G versus Cyclic Frequency (CTSS)

Average Line of Sands

(After Tatsuoka and Adachi , 1979)

E

o

h = hmax { 1- G/Go)

hmai(%) ec A 16.43 0.42

B 17.92 0.47

C 20.41 0.52

A~C 18.11

Cm

o 1.00 1.00

2.00 1.00

A 3.00 1.00

o A. 00 1.00

.3 .4 .5 .5 .7 .8 .9

Normalized Shear Modulus G/Go

Fig. 10 Damping Ratio versus Normalized Shear Modulus

( CTSS)

219
Fig. 11(a) Relationship between Normalized Shear Modulus G/Go and Shear Strain r

10"^ 2 5 10 ^ 2 5 10"^ 2 5 10

Shear Strain ^

Fig. 11(b) Relationship between Daiping Ratio h and Shear Strain r

220
Allowable Residual Displacement of Gravity Quaywall Given by

Optimum Seismic Coefficient From Economical Viewpoint

By

Tatsuo UWABE^

SUMMARY

Cases of earthquake damage to gravity

quaywalls were collected for past

earthquakes, and the quantity of the

earthquake damage which means the

residual displacement after earthquakes and the cost of damage was

then analyzed. An estimation method

to give this quantity of the damage

(residual displacement and cost) was presented, using the empirical equation which is the function of the

ratio of the corresponding seismic

coefficient of the ground acceleration to the seismic coefficient which

gives the safety factor of unity in

the stability analysis of the design

standard. An optimum seismic coefficient from an economical viewpoint

and an allowable residual displacement which was defined as an expected seismic damage displacement given

by such optimum seismic coefficient

were then studied for the rational

seismic design, on the basis of the

quantitative estimation method of the

cost of damage.

Keywords : earthquake resistant design,

seismic damage, residual displacement,

seismic coefficient, gravity quaywall

1

.

INTRODUCTION

Currently in Japan, there is a high

possibility of the occurrence of a large earthquake in Tokai Area, a central part of Japan, in the near

future, and thus many kinds of investigations for earthquake preparedness have been done. One of the

investigations related to ports that

may become key locations for the

transportation of emergency goods

immediately after an earthquake, is a survey on the earthquake resistance

capability of port facilities to

estimate the number of port facilities available after the earthquake.

In this survey, the seismic stability

of port facilities has been judged by

the evaluation method reported by

Tsuchida et al-'-^. According to this

method there are only the two kinds

of analyzed results of no damage, and

collapse. However the actual seismic

damage of port facilities shows a gradual change from no damage to

collapse

.

It is now necessary to assess the

potential seismic damage quantitatively by the following reasons.

i) If the extent of damage to port

facilities is small, they can still

serve for the temporary transportation of urgent goods. It is therefore important to know whether the

port facilities are available

after earthquakes, by prior estimation of the extent of damage.

ii) Because the number of port facilities which are assessed to be damaged is large and because they

cannot be reinforced at the same

time, it is necessary to decide the

priority of reinforcement and an effective reinforcing method ac- cording to the extent of potential

damage

.

With the background described above,

the present study is concerned with

the development of a quantitative

estimation method of seismic damage

to port facilities. It is further

aimed at establishing the rational

earthquake resistant design by utilizing the quantitative estimation

method of seismic damage thus developed. In this earthquake resistant

design, it is necessary to give an allowable value of seismic damage

deformation. The seismic damage

deformation means the swelling and

settlement of face line of the wharf,

the tilting of wall and so on. The

allowable value is considered to be

given from the viewpoints of the

berthing function, the structural

stability, the economy and so on. In

the present study, the allowable

residual displacement from the eco- nomical viewpoint is investigated. An

optimum seismic coefficient from the

•Dr. Eng., Chief of Earthquake Disaster Prevention Laboratory, Structural

Engineering Division, Port and Harbour Research Institute, Ministry of

Transpor t

221
economical viewpoint and an allowable

seismic damage given by such seismic

coefficient is sought for, using the

quantitatively estimated potential

seismic damage. The structures ana- lyzed in this study are a gravity

type quaywall, which is a typical

berthing facility in Japan, and a gravity type revetment.

In this report, historical cases of

seismic damage to gravity quaywalls

are first described. On the basis of

these cases, a quantitative estimation method of seismic damage to

gravity quaywalls is then presented.

Lastly, an optimum seismic coefficient and an allowable residual

displacement are derived by applying

the quantitative estimation method to

several cases of gravity structures

with the cost analyses.

2. CASES OF SEISMIC DAMAGE TO GRAVITY

QUAYWALLS

2 . 1 Earthquake, Ports and Gravity

Quaywalls

A gravity quaywall is a typical

berthing facility to moor ships in

Japan. The structural types of gravity quaywalls are classified into the

caisson type, the concrete block

type, the L-shaped concrete block

type, the cellular block type and the

wave absorbing vertical wall type.

Figure 1 illustrates the caisson type

of gravity quaywalls. Table 1 shows

the number of port facilities damaged

by past earthquakes from 1923 to

1978. The details of these 679 data were shown in the report of Uwabe^^.

In the study on the quantitative

estimation of seismic damage to

gravity quaywalls the episodes of

gravity quaywalls clearly damaged by

liquefaction were excluded from this

report

.

2 . 2 Quantification of Damage to

Gravity Quaywalls

(1) Seismic Damage Deformation

In a survey of seismic damage the

swelling and settlement of face line

of the wharf, the tilting of wall,

the settlement of apron and other

factors as shown in Fig. 2 was measured, and the length of damaged sec- tion in one berth (damage length)

was also measured. In this report,

these measured values and the damage

deformation ratio( the ratio of the

maximum swelling to the wall height

which is the height between sea bed

and wall head) were used as the

parameters to quantity the damage

def ormat ion

.

( 2 ) Cost of Seismic Damage to Gravity

Quaywalls

The cost of seismic damage to gravity

quaywalls (seismic cost) means the

outlay assessed officially as repair

work by the government. In this

report, the seismic cost was defined

as the sum of the repair work cost

divided by the damage length of

gravity quaywall (unit: 1,000 yen/m).

In this study, the ratio of the

seismic cost to the initial construction cost was also discussed. This

ratio is termed the cost rate of

seismic damage (seismic cost rate).

The costs shown in the earthquake

damage reports are the sums of the

day. It was therefore required to

convert these costs to same price

level. Then, the fluctuation of the

past years was investigated in the

construction prices, in the wholesale

prices of construction materials and

in the wages respectively, and the

fluctuation was quantified by a price

index that is 100 for the prices at

the year 1980. The repair cost and

the initial construction cost were converted to same price level by

this price index^ ^

.

3. QUANTITATIVE ESTIMATION OF SEISMIC

DAMAGE TO GRAVITY QUAYWALL BY

ANALYSIS OF PAST DAMAGE DATA

3 . 1 Assessment of Damage Occurrence

A working seismic coefficient means

the seismic coefficient which works

on structures during earthquakes, and

the relation between the working

seismic coefficient and the maximum

ground acceleration was presented as follows-^ ^ .

1/1

(a<200 Gal)

(0^200 Gal)

(1)

where Ke :Working seismic coefficient

Q :Maximum ground acceleration

(Gal ) q : Acceleration of gravity

(980 Gal )

222
For gravity quaywalls, the stability

analysis should aim to derive the

sliding and overturning behaviors as well as the bearing strength of the

foundation, and also determine the

seismic coefficient causing each of

these failures under the safety

factor of 1. This seismic coefficient

is termed the critical seismic coefficient. Critical seismic coefficients are given respectively in

three stability examinations mentioned above. When the smallest of

these critical seismic coefficients

is smaller than the working seismic

coefficient, the structure starts to

break during earthquakes. This smallest critical seismic coefficient is

defined as the breaking seismic

coef f ic lent

.

A decision on whether or not damage

would occur is based on the comparison between the working seismic

coefficient and the breaking seismic

coefficient. If the breaking seismic

coefficient is greater than the

working seismic coefficient, the

structure is considered safe and

earthquake-resistant. Otherwise, the

structure is expected to sustain

damage due to the earthquake.

3 . 2 Quantitative Estimation of Seismic Damage to Gravity Quaywall

As mentioned in the last paragraph

the structure starts to breaking when

the working seismic coef f ic lent ( Ke

)

is greater than the breaking seismic

coef f ic lent ( Kc )

, and the quantitative

damage that occurs to a structure is

considered proportional to the ratio

between Ke and Kc . Therefore, a method to estimate the damage extent was presented in this study using

this ratio between Ke and Kc . The

ratio of Ke to Kc was defined as the

risk ratio(Fc). The relationship

between the seismic damage deformation and the risk ratio was then

investigated on the basis of the

regression analysis of the historical seismic damage data. In this

analysis, the maximum swelling, the

settlement of the face line, the

damage deformation ratio and the

seismic cost rate were discussed in

this report.

Figure 3 shows the relationship

between seismic damage deformation

ratio (Rg) and the risk ratio (Fc).

This figure shows the extent of

damage distinguished by the symbols.

This extent of damage was classified

into five categories between no damage and complete collapse^^.

Table 2 shows the equations representing regressions. The regression

formula of Rg and Fc in Fig. 3 was

then given as follows.

Rg = -12.7 + 14. 5Fc (2)

As shown in Table 2, correlation

coefficients of the regression formula obtained here are not thoroughly

high. In order to obtain a high

accuracy, it is necessary to reexamine the relationship between the

working seismic coefficient and the

ground acceleration with higher

accuracy, to investigate the relationship between the seismic damage

deformation and the risk ratio for

each failure mode, considering other

factors of the ground condition and

so on. However, no one knows whether

a large number of damage data for

regressions with a higher accuracy

will be obtained or not in the near

future. Therefore, it may be proper

in the present situation to use the

regression formula obtained here for

this study.

The regression formula that represents the relationship between the

seismic cost rate (Cf) of the gravity

quaywalls and the risk ratio (Fc) was not obtained with high accuracy

because of insufficient number of

data covering the cost of seismic

damage and result of stability analysis to give the breaking seismic

coefficient. Then, the relation between Cf and Fc was presented from

the two regression formulae that are the equations of Cf and the damage

deformation ratio (Rg) in Fig. 4, and

that of Rg and Fc in Fig. 3. According

to these two formulae, the relationship between Cf and Fc can be given

as f ol lows

.

Cf=-62. 2+66. 2Fc (3)

It was believed that this formula was obtained by means of the best method

in the present situation in order to

estimate the cost of gravity quaywalls for an optimum seismic coefficient from the economical viewpoint

discussed in next chapter.

223
4. OPTIMUM SEISMIC COEFFICIENT AND

ALLOWABLE RESIDUAL DISPLACEMENT

FROM ECONOMICAL VIEWPOINT

4 . 1 Concept of Optimum Seismic CoeffTcienf^)^) and Allowable Residu~

al Displacement

One of the seismic coefficient given

by the rational earthquake resistant

design method is the optimum seismic

coefficient from the economical

viewpoint. The factors to affect the

definition of an optimum seismic

coefficient from an economical viewpoint are the initial construction

cost, the cost of seismic damage to

the structures, the utility of

repair works to local economics, the

effect of port constructions on the

environmental and so on. At the

present time, it is very difficult to

quantify these factors other than the

initial construction cost and the

cost of seismic damage to structures.

Therefore the economical viewpoint in

this report focused on these two

factors for the first step to define

the optimum seismic coefficient.

When the seismic coefficient becomes

larger, the expected cost of seismic

damage to structures (expected seismic cost:C£(k)) decreases, and the

initial construction cost(Ic(k)) increases, as shown in Fig. 5. Therefore, it is believed that the sum of

the initial construction cost and the

expected seismic cost (expected total

cost:Ct(k)) shows a raised down

curves with the extreme. In this

report, the seismic coefficient which

gives the extreme of this Ct(k) is

defined as an optimum seismic coefficient from an economical viewpoint.

And the seismic coefficient in this

study is assumed to be the same as the breaking seismic coefficient

because the seismic coefficient of

the present design method for the

gravity quaywall is nearly equal to

the breaking seismic coefficient.

An allowable residual displacement

for the earthquake resistant design

from the economical viewpoint was investigated in this paper as described in Chapter 1. The allowable

residual displacement was defined as the expected seismic damage displacement given by the optimum seismic

coefficient mentioned in the last

paragraph

.

The accuracy of estimating the expected seismic costs discussed here

is not necessarily very high, considering the scattering of the expected

seismic costs and that of the acceleration for the attenuation curves

with distance. However, the study on the quantitative estimation of the

expected seismic cost is the first

stage in earthquake engineering. It was still considered for this study

to be currently useful.

4 . 2 Relation between Initial Construction Cost and Seismic Coefficient

The initial construction cost of a given gravity revetment which was designed for the two kinds of foundation ground of a sand layer and a clay layer, and for several seismic

coefficients was estimated^^. Figure

6 shows this initial construction

cost versus seismic coefficient. The

vertical axis in Fig. 6 is the ratio

of the initial construction cost for

each seismic coefficient to that for

the ordinary condition.

The relationship between the initial

construction cost and the seismic

coefficient for port facilities had

been reported in the study on the

economical design of port facilities

by Murata, Yagyu and Uchida^^. Figure

7 shows the initial construction cost

against the seismic coefficient for

the gravity quaywall where the

structure is of the caisson type

shown in Fig. 8. It is necessary to

pay attention to the price level of

the year 1976 and the unit of vertical axis that is 10,000yen/m.

4 . 3 Expected Seismic Cost Rate

( 1 ) Cumulat ive Distribution Function

and Probability Density Function

of Maximum Ground Acceleration

at Japanese Ports

The probability of the occurrence of

the maximum ground acceleration at a given site was studied here based on

the report of Kitazawa, Uwabe and

Higaki'^. Figure 9 shows the base

rock acceleration against the return

period at Tokyo, Niigata and Shimonoseki. These three ports were se- lected from the viewpoint of probability of earthquake occurrence.

According to Fig. 9, the maximum base

rock accelerations for the return

224
period of 50 years are about 240 Gal

at Tokyo port, about 120 Gal at

Niigata port and about 60 Gal at

Shimonoseki port.

The cumulative distribution function

and probability density function of t durable years were obtained as follows^'^)

.

F,(;^J=[(l-Y)+Y-exp(-m-e-^)

t-N

(m-D! A-K

(n-D!

exp(

')!' X„-B

m z TTi' e (5)

where, Xm : m-th acceleration

A,B : Constant of Gumbel

distribution

N : Number of data

K : Period of earthquake

data

t : Durable years

m : Order of extreme

Figure 10 shows the distribution

function and the probability density

function for Tokyo port. The probability density function of the maxi- mum anticipated acceleration at Tokyo

port for the durable years of 50

years is f]^(x) in Fig. 10, and the

acceleration of about 240 Gal where

f]^(x) shows a peak is the same as the

expected acceleration for the return

period of 50 years in Fig. 9.

(2) Expected Seismic Cost Rate

When the distribution function of the

m-th extreme is fm{x), the probability of occurrence of the m-th extreme

is fm(x)dx. The expected seismic cost

rate is derived from fm(x)dx and

random D(x)which is the seismic cost

rate as follows

.

Pc/=E f D{x)Ux)dx

JlX) (6)

Moreover, D(x) is given by the equation of ( 3 ) . In case of the expected

seismic damage deformation ratio,

D(x) is given by the equation of (2).

As the extreme of large order had

little influence on the expected

seismic cost rate, the extremes from

1st to 5th were considered for the

calculation of the expected seismic

cost rate. Figure 11 shows the

expected seismic cost rate of Tokyo

port for durable years of 25, 50 and

100.

4 . 4 Optimum Seismic Coefficient and

Allowable Residual Displacement

Optimum seismic coefficients to

minimize the expected total cost

were calculated for Tokyo

port, Niigata port and Shimonoseki

port. The number of durable years

was 50, and the extremes from 1st to

5th were considered. The relations

between the initial construction cost

and the seismic coefficient in Figures 6 and 7 were used. As the amount

of the initial construction cost in

Fig. 6 was not shown, it was supposed

that the initial construction cost of

the ordinary condition was 1,000,000

yen/m. As the year of the price level

in Fig. 7 is the year 1976, the amount

of the expected seismic cost was

converted in the price level of the

year 1980.

Results of calculation are shown in

Fig. 12 for the gravity revetment and

in Fig. 13 for the gravity quaywall

(Caisson type). The solid lines with

symbols of X in Figures 12 and 13

show the initial construction cost.

The solid lines with closed circles,

open circles and squares show the

expected seismic cost. The dotted

lines show the expected total cost.

Table 3 shows the optimum seismic

coefficient to minimize the expected

total cost in Figures 12 and 13.

The optimum seismic coefficients

obtained here, were compared with the

expected maximum ground acceleration

with a return period of 50 years.

Table 3 shows the seismic coefficients transformed from the expected

maximum ground accelerations for a return period of 50 years in Fig. 9,

using Eq.(l). In the case of gravity

revetments, the optimum seismic

coefficients were larger than the

seismic coefficients for a return

period of 50 years. In the case of

gravity quaywalls, the optimum seismic coefficients were the same as the

seismic coefficients at Tokyo port

and were slightly larger than those

at Niigata port and Shimonoseki port.

Figures 12 and 13 show the expected

maximum swelling calculated from

225
Eq.(2). The height of structures is 5 m for the gravity revetment and is

14m for the gravity quaywall. The

expected maximum swellings of the

optimum seismic coefficient are as follows. In the case of the gravity

revetment, the expected values of

maximum swelling were 5 cm at Tokyo

port, 10cm at Niigata port and 1 cm

at Shimonoseki port. The reason why

the expected maximum swelling at

Tokyo port is smaller than that at

Niigata port is that the increasing

rate of the initial construction cost

between 0.15 and 0.2 is larger than

that between o.2 and 0.25. In the case

of the gravity quaywall, the expected

values of the maximum swelling were

67 cm at Tokyo port, 50 cm at Niigata

port and 10 cm at Shinomoseki port.

This displacement is an allowable

residual displacement, when defined

from an economical viewpoint. In

addition to the displacement discussed here, an allowable displacement from the viewpoint of the berthing function, the structural stability and so on should be examined for

the rational earthquake resistant

des ign

.

5. CONCLUSION

Data on cases of seismic damage to

gravity quaywalls were collected.

Then the quantification of the earthquake damage and the quantitative

estimation method of seismic damage

to gravity quaywalls were investigated. The relation between the damaged

deformation ratio and the risk ratio

which is the ratio of the working

seismic coefficient to the breaking

seismic coefficient, and the relation

between the seismic cost ratio and

the risk ratio were obtained on the

basis of the seismic damage data of

gravity quaywalls in past earthquakes . Moreover, the procedure to give an optimum seismic coefficient from an economical viewpoint and the allowable residual displacement which was defined as the expected seismic

damage displacement given by such

optimum seismic coefficient was

presented, using the method for

estimating the cost of seismic

damage to gravity quaywalls. The

optimum seismic coefficients and the

allowable residual displacement of

the gravity revetments and quaywalls

were obtained from the expected total

cost with the durable period of 50

years at Tokyo port, Niigata port and

Shimonoseki port. The results of a comparison between these optimum

seismic coefficients and the working

seismic coefficient calculated from

the expected maximum ground accelerations for the return period of 50

years were as follows. In the case of

gravity revetments the optimum seismic coefficient was larger than the

working seismic coefficient with the

return period of 50 years. In the

case of gravity quaywalls the optimum

seismic coefficients were same as the

working seismic coefficient at Tokyo

port and were slightly larger than

those at Niigata port and Shimonoseki

port. The working seismic coefficient

in Table 3 are nearly equal to the

seismic coefficient used in the

present design method of gravity

quaywalls for each port. As the

seismic coefficient of the present

design method had been established

empirically on the base of the past

earthquake damage, it seems that the

result of this study on the optimum

seismic coefficient shows the validity of the empirical engineering

judgment from the economical viewpoint for the seismic coefficient.

REFERENCES

1) Tsuchida, H., Noda, S., Inatomi,

T., Uwabe, T., Yagyu, T. and Murata,

T. :Method of Evaluation for Seismic

Stability of Port and Coastal Facilities, Technical Note of the Port and

Harbour Research Ins t i tu t e ( PHRI ) , No. 336, June 1980

2) Uwabe, T . : Es t imat ion of Earthquake

Damage Deformation and Cost of Quaywalls based on Earthquake Damage

Records, Technical Note of PHRI,

No. 473, Dec. 1983

3) Noda, S., Uwabe, T. and Chiba,

T.: Relation between Seismic Coefficient and Ground Acceleration for

Gravity Quaywall, Report of PHRI,

Vol.14, No. 4, Dec. 1975, pp. 67-111

4) Nonaka, M., Inokuma, Y. and Katayama, T.:Studies on Seismic Damage

Rate and Optimum Seismic Coefficient

Determination of Highway Bridges,

Proc. of JSCE, No. 340, Dec. 1983,

pp. 87-96

5) Matsuo, M. : Reliability in Geotechnical Engineering Design, Gihodo

Press, March 1984

226
6) Murata, T., Yagyu, T. and Uchida,

T.:Some Consideration on Profitable

Designing for Port and Harbour Facilities, Proc . of 1980 Annual Research

Presentations of PHRI , Dec. 1980,

pp. 233-276

7) Kitazawa, S., Uwabe, T. and

Higaki, N.:Expected Values of Maximum

Base Rock Accelerations along Coasts

of Japan, Technical Note of PHRI,

No. 486, July 1984

8) Gumbel, E . J . : S t a t i s t i c s of Extremes (Japanese translation

version) , Seisan-Gijyutsu-Senta-Shinsya, June 1978

9) Uwabe, T. : Study on Quantitative

Estimation of Seismic Damage to

Gravity Quaywall, Technical Note of

PHRI, No. 548, June 1986
Table 1 Data of seismic damage to port facilities

Number of

Earthquakes

17

Number of

ports

100

Structural

type

Quaywall Gravity type

Steel sheet

pile type Cellular type Piled pier type

Breakwater

Other types

Total

Number of

damage data

275(77)

161(56)

IK 7)

31(17)

40(11)

161 ( 39 ) 679(207)

Figures within parentheses are no damage data

ana are included in total of each type.

Table 2 Results of regression analysis

Criterion variables

Maximum swelling

( Dx , cm

)

Settlement of face line

( Sp , cm

)

Damage deformation ratio

(Rg,%)

Regression formula

Dx=-113. 8+124. 4Fc

Sp= -50.9+ 57.1FC

Rg= -12.7+ 14.5FC

Correlat ion

coef f ic ient

0. 559

0. 667

0.445

s tandard

deviat ion

59.

1

20. 0

9.1

Predictor variable (Fc):Risk Ratio(Ke/Kc)

Table 3 Optimum seismic coefficient and allowable residual displacement

Name of

por t Optimum seismic coef.

I Revetment Quaywall

Seismic coef. lAllowable residual dis.(cn

(Return period

of 50 years) I Revetment Quaywall

Tokyo I 0.23

Niigata I 0.15

Sh imonosek i I 0.10

0. 21

0. 13

0.08

0.21

0. 12

0 . 06

5 10

1

67

50

10

228
20.0O

(Unit : m )

Fig.l Gravity Quaywall(Caisson Type)

" Swelling of Face Line

J if"" I Settlement of Apron

Settlennent:}: r--7 '/ of Face Lin^

'

Tilting

->-r 1 7—I—TT—<—1 T " 4 IT

Fig. 2 Seismic Damage Deformation

100

80

oI 60

c

o

2 40

20

Extent of

Damage 0 I X E

in

IV

Risk Ratio (Ke/Kc) Fig. 3 Damage Deformation Ratio

versus Risk Ratio

300

250

<u 200

4-> ro

o

<_> 150

100

50

Extent of

o I De

X n

A HI a IV

image

tcr ^j,.-'

— >^ ^^.^^^^

^"'^

1 1 1

1 1 1 1 ' 10 20

Damage Deformation Ratio (%)

Fig. 4 Seismic Cost Rate versus Damage Deformation Ratio

30

229
Seismic Coefficient

Fig. 5 Relation between Cost and Seismic

Coefficient

COoo

co

3

COc

o

1.5

1.0

„ 0,5 a

0

Sand layer

i.O

Gravity Revetment+ 6.0 4.000

1.15 1.17

0.1 0.15 0.2

Seismic Coefficient

025

o 2.0

o

<_> o

=3

o

1.5

1.0

05

0

Clay layer

1.0 ______

L08.„-.--^

0.1 0.15 02

Seismic Coefficient

025

Fig. 6 Initial Construction Cost versus Seismic Coefficient (Gravity Revetment)

230
500r

o

oo

400-

300 -

? 200

t: 100

0.05 0.1 0.15 0.2

Seismic Coefficient

0.25

Fig. 7 Initial Construction Cost versus Seismic Coef f icient (Murata et al.°0

Unit ; m

Mooring Post Coping

Fender

20.00

Fig. 8 Standard Section of Caisson Type Quaywall

(Murata et al . ^ h

2000

1000

500

200 ^ 100

o

S-

(1) a.c

s-

+->

50

20

I 0

5

Tokyo Port

E 01885- 1981 t

1 1 1 r 1 III 1 1 1 1 1 1 1

1

2000

1000

500

5-<o 200

<u>> v-^lOO

-oO 50

U

Q. 20 c

S-

'+-> Ol

Niigata por t E 01885-1981

= 7^ ^

1 1 1 1 II 1 1 I 1 1 M 1 II

2000

1000

500

s- 200

>> 100 o

o 50

i-dJ Q. 20 c

S- 10 3

+-> OJ 5

CC

1 n^ / -Z 1

1 1 1 M 1 II

Shomonoseki

Port

OI885- 1981

1 1 M 1 Ml

10 20 50 too 200 500 1000 Acceleration (Gal)

10 20 50 100 200 500 1000 Acceleration (Gal)

10 20 50 100 200 500 1000 Acceleration (Gal)

Fig. 9 Maximum Base Rock Acceleration versus Return Period

231
0 100 200 300 400 500 Maximum Acceleration (Gal)

Fig. 10 Distribution Function and Probability

Density Func t ion ( Tokyo Port)

Seismic Coefficient

Fig. 11 E.xpected Seismic Cost Rate versus Seismic Coefficient

232
Fig. 12 Expected Total Cost versus Seismic

Coefficient (Gravity Revetment)

^ Initiol Conslruclion Cojl *,__JiEKp«cled Stiimic Coil 8000 • 3U.p„led S-.llmg

Grovtty OuoywolKCoiSSOn Type) \ ^^"-i; E,p«cled Toiol Coti

•• Tokyo PofI Dufoble Yeors '. SOyeors I

! >J Nrigolo Port

IL' Shimonstkc Porlj

Fig. 13 E.xpected Total Cost versus Seismic

Coef f icient(Gravity Quaywall)

233
The Relation Between the Building Damages and the Vibration

Properties of Ground (In the Case of Spitak Earthquake)

By

Keiichi OHTANI''', Hiroyoshi KOBAYASHI

SUMMARY

The tragic earthquake of December 7,

1988 (Spitak earthquake) occurred near

Spitak, Armenia, USSR. The earthquake

resulted severe damages of buildings at

the northern part of Armenia. And that

earthquake also resulted in a loss of

life estimated at greater than 45,000.

The buildings, especially the reinforced

concrete apartment houses were suffered

very severe damages.

The causes of this disaster were mainly

estimated that (1) the very strong

ground motion was attacked to the hazard

area, and (2) the natural period of

building was consisted with the

predominant period of ground.

This paper described that the estimation

of the intensity of ground motions by

assuming a fault-rupture model and by

considering geological conditions, and

the relation between the building

damages and the dynamic properties of

ground by the data of microtremor

.

KEY WORD

Spitak earthquake. Seismic intensity.

Building damages, Dynamic behaviour of

ground, Microtremor measurement.

1 . INTRODUCTION

The tragic earthquake of December 7,

1988 occurred near Spitak, .^^rmenia,

USSR. The earthquake resulted severe damages of structures, especially the

apartment housing buildings, at

Leninakan, Spitak, Kirovakan, Stepanavan

in the north-western part of Armenia.

And, this earthquake also resulted in a loss of life estimated at greater than

45,000; the actual death toll may be

significantly higher. The damages of

structure were concentra te 1 y occurred to

the buildings, especially the apartment

houses of reinforced concrete

structures

.

The lessons learned from past damaging

earthquakes have pointed out that there

is an evident relation between

earthquake damages and geological

conditions, because the intensity and

characteristics of ground motions depend

on geological conditions.

The objectives of this paper are to

estimate the intensity of ground motions

due to Spitak Earthquake by assuming a fault-rupture model and by considering

geological conditions and to discuss the

relations between their results and the causes of damages due to Spitak

Earthquake by using the microtremor data

on buildings and ground surfaces.

2 . ESTIMATION OF SEISMIC INTENSITY

2.1 Parameters of earthquake and

distribution of after shocks

The parameters of Spitak earthquake as reported by the Soviet Academy of

Science are : Magnitude; M=6.8 - 7.0

Origin time; December 7, 1988

llhr41min , local time

07hr4lmin (GMT)

Latitude ( deg ) ; 40. 9N

Longitude ( deg ) ; 44. 2E

Depth; 10km

Based on the recorded data at Earthquake

Research Institute, Armenia Academy of

Science, the fault plane solution is

shown in Fig. 1(c). This solution was

the reverse fault with right lateral

component, the strike and dip angle of

the fault were estimated to N60 W, 60 , respectively

.

The Figs. 1(a) and 1(b) which were measured by Armenia Earthquake Research

Institute are shown the distributions of

aftershocks. The Fig. 1(a) is shown the

distribution of aftershocks which were measured within the duration between

December 7, 1988 and January 15, 1989,

and the Fig. 1(b) is shown the

distribution of aftershocks within

December 7, 1988, because the shape of

fault corresponds to the distribution of

aftershocks measured within 24 hours

after a mainshock. The center and

radius of circles in those figures are shown the location and magnitude of

earthquakes, respectively. From these

figures of distribution of aftershocks,

the shape of earthquake fault was

*1 Head, Earthquake Disaster Prevention

Laboratory, National Research Center for

Disaster Prevention, Science and

Technology Agency

*2 Professor Emeritus, Tokyo Institute

of Technology

234
estimated as follows:

Length; L=35km, Width; W=12km

Dip angle; 60 , Strike; N60 W. These values are coincided with the

values by Armenia Earthquake Research

Institute

.

The fault slips were discovered in the

hill area on the west of Spitak. The

maximum vertical and horizontal

components of the fault slips which were discovered, are the value of 2.0 m and

1.8 m, respectively.

2.2 Estimation of seismic moment

The seismic moment of 2^^^ earthquake is

the value of 2.5X10 dyne cm which

is calculated by the parameters of

L=35km, W=12km and D=l . 5m (mean value).

By using this seismic moment, it can be

introduced to the magnitude of M=6.8

(Table 1). The parameter of D=1.5m is

chosen as the mean value based upon the

discovered fault slips.

Fig. 2 shows the location and parameters

of the earthquake fault. The rectangle

and double-lined part in this figure are illustrated to the location and scale

and the shallow part of the fault,

respectively. And, the black circle is

illustrated to the epicenter.

2 . 3 Maximum Accelerations and Seismic

Intensity

The maximum accelerations at each place

are calculated as the product ' of those

of the incident waves from the seismic

bedrock and amplification factors by

using the jripthod of Drs . Midorikawa and

Kobayashi. We assumed the rupture

velocity {V ) is 2.5 km/sec by

experience. The surface-soil conditions

at each place are incorporated on the

basis of Armenia geological map

(1/6,000,000) as shown in Fig. 3. The

amplification factors to the seismic

bedrock at each place are shown in Table

2, considering the surface geological

conditions. The seismic intensity (MSK) are evaluated on the basis of the

maximum accelerations which are reduced

to about 80% levels of the calculated

ones. The reduction factor depends on

the experimental relation between

maximum accelerations and seismic

intensity in Japan. The results are shown in Table 3 . The maximum acceleration level of Spitak

2 in Table 3 is expressed as more than

600gal, as the past results show good

agreement in the range of maximum

accelerations less than 600gal or so, in case the calculation is carried, out in

the elastic manner adopted here. The

maximum acceleration level of Gukasyan

is expressed as more than 228gal, as the

geological condition is assumed as hard

rock because of the insufficient

information

.

2 . 4 Appraisal of the seismic intensity

The relations between the calculated

results of seismic intensity by the

assumption of fault-rupture model and

the reported seismic intensity and the

earthquake damages are summarized as follows

:

(a) The maximum acceleration levels of

Gukasyan and Yerevan agree with those of

available accelerograms recorded in

Gukasyan and Yerevan which are about

210gal and 61gal, respectively.

(b) The seismic intensity are 10 in

Spitak, 9-10 in Leninakan, 8-9 in

Kirovakan, 6 in Yerevan in terms of MSK

scale and agree with the estimated ones

in Armenia.

(c) The results are supposed to explain

that the intensity of ground motions in

Kirovakan more close to the epicenter is

lower than that in Leninakan far from

the epicenter, although the final

judgement requires the exact seismic source mechanism and geological

condition

.

2 . 5 Comments on the Seismic Zoning Map

for the seismic design

The seismic zoning map for the seismic

design is used generally in Armenia.

But, the method of seismic zoning map,

which is adopted in Soviet, is similar

to based upon the method of seismic macro zoning map in Japan. Therefore,

the method in Soviet does not considered

the method of the micro zoning, namely

based upon the dynamic properties of

ground. It is seemed that the method of macro zoning map is mainly based upon

the statistical result of the occurrence

of historical earthquakes, and is not

considered the Sei smotectonics , which is

strongly insisted by the Soviet Academy

of Science in former times.

This Armenia region is located the very

active seismic zone, where the Anatoria

fault is closed to the west side of

Armenia and the Zakros seismic belt is

connected to the south-eastern direction

of Armenia through the Azerbaizian. The

present zoning map of Armenia is

difficult to understand for the

considering with the face of geological

structures

.

We hope that the seismic zoning map will

235
be corrected based upon the results of

the distribution of the seismic

intensity of each city at this

earthquake and the seismotectonics of

these area

.

3. DYNAMIC PROPERTIES OF BUILDINGS

3 . 1 Microtremor measured on buildings

The buildings which were suffered severe damages were concentrated to the

uniformly produced reinforced concrete

apartment houses. Therefore, the

microtremor measurement on buildings

were conducted to the research for the

change of dynamic behaviour between the

undamaged and slightly damaged buildings

and the relation of the occurrence of

damages. For the measurement of

undamaged buildings, we conducted in

Yerevan, the capital of Armenia. For

the measurement of slightly damaged

buildings, we conducted in Leninakan and

Kirovakan

.

The structural typ'es of measured

.buildings were selected to the 9-story

precast reinforced concrete frame

buildings, 9-story large-size panel

buildings, 5-story stone-masonry

buildings with reinforced frame and 4 and 16-story site-casted reinforced

concrete buildings.

3 . 2 Natural periods and critical dampin g coefficients of building ~

The microtremor on the buildings were measured 19 sites in Yerevan, Leninakan

and Kirovakan. The measurement in

Yerevan were recognized for the dynamic

properties of undamaged buildings. The

measurements in Leninakan and Kirovakan

were recognized for the properties of

slightly damaged buildings. The

measurement results are shown in Table

4 . The change of dynamic properties between

damaged and undamaged buildings for 3 structural types are appointed out as

f ol lows

:

1) 5-story Stone-Masonry Building : The

natural periods and critical damping

coefficients of undamaged buildings are

0.28 - 0.33 sec. and 2.2 - 7.2 %,

respectively. The natural periods and

critical damping coefficients of damaged

buildings are 0.21 - 0.55 sec., 1.9 - 4.4 %, respectively. The difference of

the natural periods of damaged and

undamaged buildings are so large,

because it is guessed that this causes

is based upon the deteriorative effects

of wal 1 s .

2) 9 -story Precast Reinforced Concrete

Frame building : The natural periods of

undamaged buildings in the longitudinal

direction and transverse direction are

0.61 - 0.6 3 sec. and 0.4 6 - 0.5 3 sec,

respectively. The natural periods of

damaged buildings in the longitudinal

and transverse directions are 0.83

1.07 sec, 0.81 - 1.00 sec,

respectively. The natural periods of

damaged buildings are increased about 3 0

4 0 % compared with the periods of

undamaged buildings. The critical

damping coefficients of damaged

buildings are 25 % larger than those of

undamaged buildings.

3) 9-story Large-Size Panel Building : The !iatural periods and critical damping

coefficients of undamaged buildings are

0.38 - 0.39 sec and 1.2 - 2.5 %,

respectively. The natural periods and

critical damping coefficients of damaged

buildings are 0.40 - 0.46 sec. and 2.6 - 4.5 %, respectively. The natural

periods are not so changed between

undamaged and damaged buildings, but the

critical damping coefficients of damaged

buildings are almost 2 times those of

undamaged buildings.

3 . 3 Relation between the building

damages and the dynamic behaviour of

buTldings

Every structural types of building were suffered heavy damages during this

earthquake. The degree of building

damages v.;as decided by the relation

between the dynamic properties of

building and ones of ground, instead of

the difference of structural systems.

However, by the inspection of damaged

feature, the most damaged buildings were constructed by the uniformly procedure,

and had the almost same degree of

aseismic capacities. Then, it is

supposed that the building damages were concentrately occurred by the passage

the some level of strong earthquake

motion

.

For the construction of uniformly

housing complex, it is important that

the buildings have the sufficient

aseismic capacities and the wide

resistant performance to earthquake by

considering with the dynamic property of

grov:nd. We considered that this

earthquake hazard is seemed to the one example of v.'arning to the building which

is constructed by the standard design.

4 . DYNAMIC BEHAVIOUR OF GROUND AND

RELATION OF BUI LDING DAMAGES

4,1 .Measurements o f microtremor on ground surface

236
The dynamic properties of surface ground

are well known to effect the damage

degree of buildings during earthquake.

And also, the shapes of Fourier spectrum

of microtremor are well known to similar

to the shapes of Fourier spectrum of

strong motion earthquake. The amplitude

of Fourier spectrum of strong motion

earthquake is equivalent to the

amplitude of Fourier spectrum of

microtremor

.

The measurement of microtremor on the

ground surface are done for the grasping

of the dynamic properties of ground.

The measurements were made to the

following two objectives:

1) The damaged buildings have the

natural periods of around 0.5 sec. of

precast reinforced concrete frame

buildings in Leninakan and the natural

periods of around 0.3 sec. of

stone-masonry buildings in Spitak and

Kirovakan. Therefore, the microtremor

were measured by the short period

seismometer (the natural period of

pendulum is 1 sec . )

.

2) The microtremor by the long period

seismometer (the natural period of

pendulum is 3 sec . ) were measured for

the estimation of deeper ground

structures in Leninakan.

4 . 2 Microtremor measured in and around

Leninakan

The microtremor of ground were measured

at 12 points in and around Leninakan.

The measured points are shown in Fig. 4. The wave forms of microtremor are shown

in Fig. 5. Fig. 5(a) shows the six

points (11, 1, 3, 6, 9, 10) which are lined from north to south direction of

Leninakan city. Figs. 5(b), 5(c) and

5(d) show the 5 points (11, 12, 4, 7,

10) from north to south of different

line of Fig. 5(a), 4 points (2, 1, 12,

5) and 3 points (6, 7, 8) from west to

east direction, respectively. The

Fourier spectrum of each point are shown

in Fig. 6(a) and Fig. 6(b).

The relation between mean period of

microtremor and degree of building

damages in Leninakan was summarized as follows

:

1) In the northern area of Leninakan

(north part from Lenin square): The mean period of microtremor of ground is

around 0.5 sec, and this period is

close to the natural period of 9-story

precast reinforced concrete frame

buildings. So it is guessed that this

fact might be one of causes of making

damages of this kind of buildings more seriously

.

2) In the central area of the city: The mean period of microtremor of ground is

0.2 - 0.3 sec., and this period is close

to the natural period of low-rise stone

masonry buildings. This fact might be one of causes of making damages of this

kind of buildings more seriously. It is

necessary to implement the more detailed

survey, because there are under ground

flows of river.

3) In the southern area of Leninakan:

The mean period is close to the period

in the northern area, but it's amplitude

is larger than of the northern area.

4 ) General trend of dynamic properties

of ground in and around Leninakan:

Leninakan city is located on the deep

sedimentary layer, and the predominant

periods of microtremor are effected by

the local sedimental condition of

surface layer. And so, the amplitude of

microtremor of some areas have

remarkably large according to the

material of surface ground at the very

shallow part. These fact might be

caused the degree of damages of

buildings.

4 . 3 Microtremor measured in and around

Spitak

The microtremor of ground were measured

at 10 points in and around Spitak. The

measured points are shown in Fig. 7.

The wave forms of microtremor of each

points are shown in Fig. 8, and the

Fourier spectrum are shown in Figs. 9(a)

and 9(b).

The relations between mean period of

microtremor and degree of building

damages in Spitak are summarized as

f ol lows

:

1) Seismic bedrock of Spitak: The

predominant period is very short (less

than 0.1 sec.) at the foot around the

hill, where a monument was constructed

at the top of this hill. The seismic

bedrock is located at the very shallow

in these area. Therefore, the dynamic

properties does not appeared in the

microtremor, and the amplitude during

the earthquake was comparatively small.

2) The predominant periods in the heavy

damaged area of Spitak are the range

between 0.2 sec. and 0.4 sec., and these

periods are close to the natural period

of destructed buildings. So it is

guessed that this fact might be one of

237
causes of making damages of this kind of

buildings more seriously. And then, the

other cause of making damages is very

strong earthquake motion itself.

3) New developing area of Spitak: The

predominant period in new developing

area of Spitak is almost close to that

in the damaged area of Spitak.

Therefore, it is necessary to construct

the more reinforced and more stiffened

bui Idings

.

4 . A Microtremor measured in and around

Kirovakan

The microtremor of ground were measured

at 5 points in and around Kirovakan. The

measured points are shown in Fig. 10.

The wave forms and the Fourier spectrum

of microtremor of each point are shown

in Fig. 11 and Fig. 12, respectively.

The relations between mean period of

microtremor and degree of building

damages in Kiravakan are summarized as

f ol lows

:

1) The amplitude of microtremor at the

just front of the City-Office is about

1/3 times of the amplitude in heavy

damaged area. The predominant period in

heavy damaged area, where is located

very close to the City-Office, is 0.2 - 0.4 sec. This fact might be one of causes of making damages of buildings

more seriously.

2) New developing area of Kiravakan:

The predominant period in new developing

area of Kirovakan is a little longer

than the period in the damaged area of

Kirovakan .

4 . 5 Relation between the ground

properties at the narrow valley and the

earthquake damages

At the city area, which is located at

the comparatively narrow valley in the

mountain zone (for example, the area of

Spitak or Kirovakan), the relations

between the ground properties and the

earthquake damages will be generally

appointed as follows;

1) At the area of outcrop part of

seismic bedrock or on the very shallow

sedimentary layer, the damages of

buildings were not so severe. The

amplitude of microtrem.or is small at the

range of short periods, but is

relatively large at the range of longer

periods. {example: the hill area of

Spitak, point 7. the central area of

Kirovakan, point 2)

2) The building damages were more severe

by the more thickness of the sedimentary

layer. But, it is very clear to find

the relation betv;een the damage and the

selectivity of predominant period. The

amplitude of microtremor is remarkably

large at the range of short period, but

is not so remarkable at the range of

longer period. (example: the housing

complex area of northern part of Soitak,

point 2. the housing complex area of

north-western part of Leninakan, point

1 )

4 . 6 Properties of longer period

microtremor in Leninakan

The microtremor by long oeriod

seismometer (the natural period of

pendulum is 3 second) in the new developing area of northern Leninakan

find the predominant period of 3 - 4 sec. The 3 measured points, as illustrated Al , A2 and A3, are shown in

Fig. 4. The wave forms of longer period

microtremor at each point are shown in

Fig. 13.

This long predominant period is caused

by the very deep sedimentary layer on the seismic bedrock. This fact shows

that the seismic bedrock of Leninakan

area will be located at the several

kilometres bellow the ground surface.

The long predominant period in the new developing area of southern Leninakan is

close to the period in the northern part

area. Therefore, the seismic bedrock in

and around Leninakan area is located at

almost same depth.

4 . 7 Relation between the earthquake

damages and the seismic microzoning

The one of the causes of heavy damage of

buildings is able to explain the

coincidence between the predominant

period of ground and the natural period

of building. The other cause is very

strong earthquake motion itself.

Therefore, it is necessary to more consideration of dynamic properties of

ground for the Seismic Microzoning. The

measurements of microtremor on ground

and building are very effective

procedure to estimate the predominant

period of ground and the natural period

of building, respectively.

ACKNOWLEDGMENT

Measuring of the microtremor were carried out by the apparatus of Tokyo

Institute of Technology and great

cooperation on the digitization and

analysis were obtained from Dr. Kazuoh

SEO, Associated Professor and Mr.

Takanori SAMANO, Assistant, Tokyo

238
Institute of Technology. The

cooperation are deeply acknowledged.

REFERENCE

1) S. MIDORIKAW.^ and H. KOBAYASHI ; "Isoseismic Map in Near-Field with

regard to Fault Rupture and Site

Geological Conditions" Proceedings of

7th World Conference on Earthquake

Engineering, Istanbul, Turkey, Sept.,

1980 .

239
Table 1 Parameters of Fault-Rupture Model

Fault Dioension L-35 ka

D=1.5 n (mean value)

Dip Direction K60'V

Dip Angle 60'

Rupture Velocity 2.5 km/sec

Table 2 Geological Conditions and Amplification Factors

Geological conditions Amplification Factors

Thick Sediiaent 4.5

Hard Sediiaent 3.5

Tertiary Period 2.5

Paleogene Period 2.0

Table 3 Maximum Acceleration and Seismic Intensity at Each Place

Place Nane Haxiciuni AccelerationSeisndc Intensity

(HSK)

Remarks

SPITAK 2 More than 600 gal 10 Sediment(Deposit)

LENINAKAN 1 530 gal 10 Northern Area

SPITAK 1 520 gal 10 Hard rock

LENINAKAN 2 470 gal 9 Southern Area

STEPANAVAN 460 gal 9

APARAN 370 gal 9

KIROVAKAN 2 360 gal 9 Sedinient(Deposit)

KIROVAKAN 1 250 gal 8 Central Area

GUKASYAN More than 228 gal 8 Unidentified Geology

YEREVAN 65 gal 6

240
Table 4 Dynamic Properties of Damaged and Un-damaged Buildings

Structure

and

K'o. of

Story

Damage Translation Torsiona

1

Longitudinal Transverse Longitudinal Transverse

T sec h % T sec h % T sec h % T sec h %

Stone

Masonry

5

U

U

D

D

D

D

0. 30

0. 28

0.49

0.44

nU. ^7DZ

0. 55

4.8

2.2

4.0

2.6

A n 4. u 3.4

0. 33

0. 28

0. 28

0.21

n C.0

U. jZ

0. 55

7.2

4. 6

1. 9

1. 9

7 A 4.4

0. 35

0. 23

0.41

0.40

U. DO

0. 56

2.5

1.8

4.7

3.5

0. Z

5.9

0. 35

0. 23

0.41

0.40

nU. DO

0.59

4. 6

I. 1

3. 6

4.5

A 1 4. 1 4.7

Precast

r rajT>es

9

UU

u. u U

uD

nuD

D

0. 57

n

0.51

0. 53

n OQ

1. 07

1. 00

n 9,7 U. 00

1.5

11.1

3.2

0.5 —

0.79

L). DO

0.46

0. 53

U. 30

0. 96

0. 81

1 nn 1 . UU

1. 1

1. A

1.7

2.3 —

0.50

n A 1 L). 41

0.41

0. 55

n R7 U. 0 /

0. 95

0. 80

n 77 U. / (

9.3

0 0

0. 6

3.4 —

0. 54

n A 1 U. 4

1

0.42

0. 55

nU. 00

0.95

0.77

nU. 7R/ 0

15. 0

A 7

4. / 2.5

3.4 — ,

Large

Pnel

Qo

U

U

11U

0. 38

0. 38

0.45

2. 2

1. 8 4.5

0.38

0. 39

0.40

1. 2

2. 5

2. 5

0. 30

0. 29

0.46

2.3

1.4

3.4

0. 30

0.29

0.45

0.5

0. 5

1. 8

Mono] i thic

4

u 0.51 2. 6 0. 32 2. 2 0. 27 0.9 0. 27 2. 9

Mono] i thi c ]5

u 1. 10 7. 8 1. 00 6. 1 0. 96 1. 9 1. 18 1.3 — unknov>Ti U:Undamaged D:Damaged U U:Undamaged and Under Construction

241
E44'

O o

LENINAKAN

STEPANAVAN

N41

o

o o

(a)

O

o O O

O SPITAK fS) 5)

O

KIROVAKAN

o

AFTER SHOCKS Dec. 7 1988 - Jan. 15 1989

E44

STEPANAVAN

N41

O O

LENINAKAN

QD

O O

Qd

O

SPITAK <

(b)

o o

o

o

KIROVAKAN

AFTER SHOCKS Dec. 7 1988

(c)

Fig. 1 Distribution of After Shocks and Source Mechanism

242
SPITAK USSR EQ Dec. 7 1988

\ rrLENii

\

/ \

Deep Part ^\y Epicenter

^~v^^Shallow Part %APARAI-I ^

ARAGATS\m . ^

Fig. 2 Location of Fault-Rupture Model

Fig. 3 Geological Map of ARMENIA

243
O O OO O O OS CO lO iTCvl O OO Ul ^ •—1 r-V—I r—I ^ <_>

•H

244
O.S2J9 micron

1.0 o 1 cron

I I I i < LENINAKAN 12 E

I I I I

1.0 bI cron

I

I I I I I LENJNAKAH 0-! E

I I I I 0.6017 nlcron

I I I I 0. ^907 ml cron

— I.Orel cron

1^

LENINAXAW 07 ZI I I I1 S

I I sec

0.S21 1 ml cron

1.0 nlcron till LEHIHAKAW 10 E

1.0 ril cron till I I I 0. 7S6 1 ml cron

I I I I I I 10

I I

I I

1 5 »e£

Fig. 5(b) V-7ave Forms of Microtremor in LENINAKAN

LENIHAKaH 02 E

0.6 ci I c r 0 n

0.3699 ralcron

f— o . 0 s» 1 c r o n

I I

I I LENIUAKAM 01 E 0.6 ol cron

I I I I I I I 0.5309 micron

f "{

I I I I l" I I I I I I LEHINAKAN 12 E

0.6 ra 1 cro

n

0.6017 Blcron

LENIMAKAM 05 E

r— 0.6 a 1 crp

n

I I I I I I I I V I 10 15 »ec

0.6212 micron

I I ^ 1 I I V I ^ I I 1 I

' 10 15 s

Fig. 5(c) Wave Forms of Microtremor in LENINAKAN

245
LENINaKAN 06 E O.TOOD ol cron

0.6 B 1 c r o n ,11111 LEMINaKaN 07 E

0.6 n 1 c r 0 n

I I I I

I 0. S2 I 1 Dl cron

1 1 I 1

I 1 LENINaKAN 08 E

— 0.6 micron

I I I I I 0. 57S3 Bl cron

I I

I I I I I I I I I I ' I 10 li »«c

Fig. 5(d) Wave Forms of Microtremor in LENINAKAN

0. 1 ll« •!

«

*t <r«« . « cc 0.1

LtMtHAKAU «4 C

-m—I—rFig. 6(a) Fourier Specta of Microtremor in LENINAKAN

246
0.1140

C . i 0

0.0)

0.01

0 . OOt —1—r 1 I I

I—r1 1 I

' I 1—I

C.I I > 3 1 } 3

O.I

0.1

LCHINAIAM 01 E

0.01

0.01

LtMlHAKAN «1 f

0. I ICI Ml

C

O.il (IC

-I 1—r—r—m—i—r i r rFig. 6 (b) Fourier Spectra of Microtremor in LENINAKAN

247
Fig. 7 Measured Points of Mictrotremor in SPITAK

0 . IC •

«

0.144} mlt O.IT 0.1147 al 9.71 • «

Fig. 9(a) Fourier Spectra of Microtremor in SPITAK

248
SPITAX 01 S 0.:06< .icron — 0-6 "Icron 11 I

I

I I I I I I I I

SPITAIC o: H 0.6H5 .Icror,

0 . 6 1 cr 0 n

'A

, I I I I I I I >

I ' I I I SPITAK 03 N 0.:/6^ .icron

0 . 6 I c ro n

, , I I I I I

I I I ;. I I

SPITAX 0^ N

f— e.SBlcrcn

0.1724 ilcron

I I I I I SPITAIC OS H 0.6 alcron

I I I I I I I I 0.1:33 •leroB III! SPITAK 06 H 0.6 I ere

n

I I I I I I I I 0. Ht7 alcron

I I I

I 0

I I I I I

1 s

SPITAX 07 H

r— 0.6 Bicron

0.0419 Bicron

SPITAK

1 011 1 1 1 N

1 1 1

0

1 07J 11 1

B 1 c r on 0.6 B 1 c ro n

SPITAX

1 Of

1 1 1 1 N

1 1 1 01 03641 1 Bicron

0. 6 B fc ro n

I I I I

I I I I I I IS ICC

SriTAK 10 N

f— 0.6 Bicron

0. I 2 I 2 Bicron

I

I I I I I I I I I I I I 10 IS >cc

Fig. 8 Wave Formes of Microtremor in SPITAK

249
-T 1—r—I—r—n—i—i

ttjrlTAl fi

•.001 V

~i 1—1—I I I

[—I—I I I I—1—r-

•.I I I 1

I 0 . «1 It

1 .01

Fig. 9(b) Fourier Spectra of Microtremor in SPITAK

250
STEBANAVAN

SPITAX

' DILIZHAN

City Office

Fig. 10 Measured Points of Microtremor in KIROVAKAN

XI rovakan

f_ O.i m\ eron

01 N 0.3507 micron

02 N

I I I I

Kl ROVAKAN

0 . X •! cron III' KIROVAKAN 03 N

0.4 I eron

I I I I

I I I I I 0.0911 micron

I I I I C.22.12 micron

I

I I I I I I I I 10 1 i <cc

KIROVAKAN O'* N

0.4 ! croft

0.3324 nlcron

I ,

I

I I I I I

1

I

'

I ' -CM 0.1989Blcron

KIROVAKAN 05 N 0.4 m\ cron

I

I I I I I I I > 10 IS sec

I I

I

I

Fig. 11 Wave Forms of Microtremor in KIROVAKAN

251
LENINAKAN Al E

0 . e a 1 c r o n

0.7-IU Blcron

I ' I I I I I I I I I

1 " I I LENINAXAN A3 E

I I I 0.7316 Blcron

Fig. 13 Wave Forms of Microtremor in LENINAKAN

(by Long Period Seismometer)

252
New Seismic Design Specifications of Highway Bridges in Japan

By

Toshio IWASAKI". Kazuhiko KAWASIIIMA^ and KInJI IIASEGAWA^

ABSTRACT

Introduced are the new Seismic Design

Specifications of Highway Bridges in Japan is- sued on February 1990. Main Items of the revision including the seismic lateral force, e- valuatlon of inertia force considering structural response. checi< of bearing capacity for

lateral load for reinforced concrete pier and

dynamic response analysis are presented.

KEY WORDS

Seismic Design, Design Method, lllgliway

Bridge

1. INTRODUCTION

Since 1985 extensive efforts have been

undertalten to establish more rational seismic

design criteria for road bridges, with

consideration of recent advantages of

earthquake engineering and lessons gained

from recent earthquakes. New specifications were were completed in 1990.

This paper outlines the new specifications

(JRA, 1990) for seismic design of road bridges

with emphasis on the stipulations revised.

2. BASIC PRrNCIPLES FOR DEVELOPING NFW

SPECIFICATIONS

First requirement on lateral force for

design of road bridge In Japan was Included In "Draft Details of Road Structure" Issued in 1926 with considerations of the damage caused

by the Kanto Earthquake in 1923. It was

stipulated in the draft details that the

maximum lateral force expected to develop at the site shall be considered In seismic design.

After experiencing significant damages during

strong earthquakes seismic regulations were

reviewed and amended several times as shown

In Table 1. "Design Specifications of Steel

Road Bridges (Draft)" were issued In 1914, and

"Design Specifications of Steel Road Bridges"

and their revised version were issued in 1934 and 1939, respectively. Although a requirement for lateral force was stipulated

in these specifications, first comprehensive

seismic design stipulations were Issued in 1971

In a separate volume exclusively for seismic

design as "Seismic Design Specifications of

Road Bridges". It was described In the

specifications that lateral force shall be

determined depending on zone. Importance and

ground condition In seismic coefficient method

and structural response shall be further

considered In the modified seismic coefficient

method. Evaluation on soil liquefaction was

firstly Incorporated In view of the damages

caused by the Nllgata Earthquake In 1964. The

1971 specifications were revised In a form of

"Fart V Seismic Design" of "Design

Specifications of Road Bridges" In 1980. A

rational evaluation method for soil

liquefaction as well as practical

countermeasures against the liquefaction was

included In the specifications.

Road bridge Is a vital component of road,

and It has to be safe enough against an earthquake so that function of road be

maintained. It can be used without loosing any

structural functions against small to moderate earthquakes with high to moderate

possibility to occur at the site. Critical

failure causing total collapse of the bridge

has to be avoided even during a significant

earthquake such as the Kanto Earthquake in 1923. Seismic design method has been gradually

Improved with consideration on damage

experiences and advantage of earthquake

engineering. Based on such improvement of

seismic design method as well as various

countermeasures against an earthquake. It Is considered that seismic damages of road

bridge have been decreasing as a total in recent years as shown in Table 2. However it is also Important to note that new damages which

have not been Identified In the past

earthquakes have been seen In recent

earthquakes, due to Improvement of seismic

design method. This obviously shows that the

countermeasures aiming only to minimize the

damages developed In the past earthquake do

not give a credit to avoid new type of

damage In the future earthquake. In

1) Director General, Public Works Research

Institute

2) Head, Earthquake Engineering Division of

Earthquake Disaster Prevention Department,

Public works Research Institute

3) Research Engineer, ditto

revising the design specifications. It is required to take account of such new type of

253
damage patterns.

In view of the understanding on tiie damage

patterns of road bridges In recent

earthquakes and advances of earthquake

engineering, the following revisions were

included in the New Seismic Design

Specifications.

1) modification on classification of ground

condition

2) unification of seismic coefficient

method and modified seismic coefficient

method Including the revision of the

seismic coefficient

3) evaluation method of Inertia force for

continuous bridges

4) Improvement on strength of sandy soils

against liquefaction

5) evaluation of bearing capacity of reinforced concrete piers for lateral force

(ductility check)

G) design ground motion for dynamic

response analysis as well as an analytical model

Table of content of the new specifications

Is the following :

Chapter 1 Chapter 2

Chapter 3

Chapter 4 Chapter 5

Chapter 6 Chapter 7

Chapter 8

General

Basic Principle of Seismic

Design

Load and Design Condition In Seismic Design

Seismic Coefficient

Bearing Capacity for Lateral

Force of Reinforced Concrete

Piers

Dynamic Response Analysis

Structural Details in Seismic

Design

Device to Expect Reduction of

Lateral Force

(Appendix)

I . References on Liquefaction

n . Examples of Classification of Ground

Condition

in. References on Design Ground Motion

rv. Example of Calculation of Natural

Period and Inertia Force

V. Reference on Bearing Capacity of

Reinforced concrete Piers for Lateral

Force

VI. Practices of Structural Details

Table 3 shows the revised part or newly

Introduced parts In the new specifications,

and Fig. 1 shows the flow-chart of seismic design.

3. CLASSIFICATION OF GROUND CONDITION

Classification of ground condition was changed from four groups to three groups as shown In Table 4 by combining the group 2 and

group 3 into same group. Tiie classification of

the ground condition is made base on the

characteristics value at the site specified as

To =

where

Tg

H,

4 EI1

Hi

(1)

I

characteristic value (sec)

thickness of i-th subsoil layer (m) shear wave velocity of i-th sublayer

(m/sec)

sublayer's number counted from

ground surface

The change was made because most of the

seismic damage of road bridge in the past

earthquake was likely to be developed on the

group 4 with the difference between the

ground group 2 and 3 being practically very

small, and because difference of ground

motion characteristics in terms of response

acceleration spectra is few between the

ground group 2 and 3. It was also taken into

account that the classification of the ground

condition based on the difference of

characteristic value of only 0.2 second Is practically quite hard in view of the fact that

shear wave velocity Is generally estimated

from an empirical formulae based on N-value of

standard penetration test.

4. SEISMIC COEFFICIENT

In the previous specifications, the seismic

coefficient method was used for the bridges

with pier height less or equal to 15 m and the

modified seismic coefficient method was used

for the bridges with pier height larger than

15 m. The modified seismic coefficient method

was originally intended to apply to bridges

with the natural period longer than 0.5 second. However, as various foundations

becomes to be adopted on wide range of soil

conditions, even bridges with pier height less

than 15 m is likely to have the natural period

longer than 0.5 second. This always caused

confusion in selecting the seismic design

method.

Therefore, the seismic coefficient method

254
and the modified seismic coefficient method

was unified, and this Is referred to as seismic

coefficient method.

In the seismic coefficient method, the

design seismic coefficient shall be determined

by Eq.(2), but no less than 0.1.

kn = Cz • Cg • Ci • Cx • kho (2) where

kn : design horizontal seismic coefficient,

kho : standard design horizontal seismic

coefficient (=0.2).

Cz : modification factor for zone (refer to Fig. 2), Cg : modification factor for ground

condition (refer to Table 5), Ci : modification factor for Importance

(refer to Table 6), and

Ct : modification factor for structural

response (refer to Table 7). For

computing Inertia force associated

with the weight of soils and dynamic

earth pressure, Ct has to be 1.0.

In Eq.(2) the modification factors Cz and Ci are the same with the value specified In the

previous specifications. The modification

factor Cg was changed associated with the

change of ground condition classification. The

modification factor Cx, wlilch represents the

difference of lateral force due to structural

response, corresponds to the modification

factor >9 In the previous specifications. The

modification factor P was formulated based on

an analysis of 44 components of strong motion

records obtained In Japan. The new

modification factor Ct was determined from an

analysis of earthquake response spectra base

on 394 components of strong motion records

obtained In Japan. Considerations that the

seismic damage was likely to occur in bridges

with short natural period were also Included.

Fig. 3 shows an comparison of the seismic

coefficient between the previous and new

specifications assuming that Cz = Ct = 1.0.

5. EVALUATION METHOD OF INEOTIA FORCE

In the previous specification, for

computing an inertia force associated with

dead weight of superstructure, which is to be

used In seismic design of substructure, the

bridge Is divided Into several structural

segments consisting of a substructure and the

part of superstructure which Is vertically

supported by the substructure considered.

Then, the Inertia force (lateral force) used

for the seismic design of substructure of the

I-th structural segment Is assumed to be

determined as

Fd, = khiRi (3) where

Fdi : Inertia force associated with dead

weight of superstructure for design

of I-th substructure

kni : seismic coefficient considered for

1-th structural segment

. . Ri : reaction force developed at I-th

substructure due to dead weight of

the part of superstructure

supported by the i-th substructure.

However as various type of bridge

structures become to be constructed, It became apparent that the Inertia force

determined by Eq.(3) assuming that the bridge

can be divided into several strucLutal

segments Is Inadequate, because tlie InerLla

force at each substructure significantly

depends on flexibility and iielgiit of the

substructure and type of the superstructure.

Therefore, more rational method to evaluate

the Inertia force induced in the

superstructure for the purpose of seismic

design of substructure was required. Tlie same demand also arises in the bridge wliere the

superstructure and the substructure are not

separated by means of bearing supports.

Therefore, in stead of the structural

segment, a definition of seismic design

structural unit, defined as the structural

unit which can be considered to develop tlie same response during an earthquake, was newly

Introduced In the new specifications. Table 8 shows examples how the seismic design

structural unit be selected for typical

combinations of structural type and direction

for evaluating the inertia force.

Natural period and the Inertia force shall

be determined as:

l)SeismIc design structural unit consisting of

a substructure and a part of superstructure

supported by the substructure

In this case, there Is essentially no

difference with the previous specifications.

Inertia force shall be determined by Eq.(4),

and the natural period shall be computed as

T = 2.01 f 6 (4) where

T : natural period, in second, of tlie seismic design structural unit

6 : lateral displacement, in meter, of

255
the substructure subjected to a lateral force equivalent with 80 %

of the dead weight of a substructure

above tlie ground surface assumed In seismic design and the dead weight

of a part of the superstructure

supported by the substructure.

It should be noted here that only the

substructure on the ground surface assumed In seismic design shall be subjected to inertia

force because tlie seismic force below the

ground surface assumed In seismic design is disregarded In seismic design. The ground

surface assumed in seismic design Is generally

taiien as the bottom plane level of the footing

In case of pile foundation. However, when the

soil layers whose soil constant Is assumed as zero due to possible soil failure such as liquefaction, the bottom of these layers Is regarded as the ground surface assumed In seismic design.

2)Selsmic design structural unit consisting of

substructures and a part of superstructure

supported by these substructures.

This Is the case newly Introduced In the new specifications. Natural period and the

Inertia force shall be evaluated In accordance

with Fig. 4, I.e.,

Dldeallze a seismic design structural unit by

a linear elastic frame model

IDapply a lateral force equivalent with the

dead weight of superstructure and

substructure above the ground surface

assumed In seismic design, and compute the

natural period as

T = 2.01 /" 6 (5)

^

^ J w(s)u(s)^ds

J w(s)u(s)ds

where

w(s) : dead weight of the seismic design

structural unit (superstructure and

substructure above the ground

surface assumed In seismic design)

at point s (tf/m)

u(s) : lateral displacement developed in the seismic design structural unit

at point s (m) when subjected to

w(s) in a direction considered in design.

IlDdetermlne the seismic coefficient kn

depending on the natural period T. lv)compute Inertia force as

Fd = k^x F (7) where

Fd : shear force (tf) / bending moment

(tfm) due to inertia force

kh : seismic coefficient

F : force developed in tlie seismic design structural unit wlien subjected

to a lateral force equivalent witli dead weight of seismic design structural unit above tiie ground surface

assumed in seismic design (LC/tfin)

For substructure supporting girder

bridges, shearing force developed at tlie center of gravity of superstructure shall be

regarded as the lateral force for seismic

design. However, when the inertia force

computed by Eq.(7) Is smaller than the inertia

force computed by Eq.(3), the latter sliall be

adopted for design. This needs some

explanations. The inertia force computed hy

Eq.{7) Is approximately proportional to the

stiffness of each substructure. This implies

that the inertia force concentrates to the

substructures with high stiffness, while reverse Is true for the substructure with low

stiffness. Inertia force talces even negative

value when the stiffness of the substructure

Is extremely small as compared with other

substrctures. However, wiien failure of the

structure such as bearing supports occurs, the contribution of each substructure for

supporting the inertia force of

superstructure could be changed from the

distribution computed by Eq.(7). Base on such

considerations, lower limit for the inertia

force evaluated by Eq.(7) was included.

G. STRENGTH OF SANDY SOIL LAYER FOR LIQUEFACTION

l"or saturated alluvial sandy layers within

20 m from the ground surface which have the

water table within 10 m from the ground

surface, and have Dso-value on the grain size

accumulation curve between 0.02 and 2.0 mm, the liquefaction potential during an earthquake has to be checked. For those soil

layers required for the check of liquefaction,

liquefaction potential shall be evaluated in terms of liquefaction resistance factor l''!.. as

Fl = R/L (8) where

Fl : liquefaction resistance factor

R : resistance of soil elements against

dynamic load

L : dynamic load induced in soil ele256
ments during an earthquake

In the previous specifications the

resistance R of soil elements against dynamic

load was evaluated from N-value of standard

penetration test and Dsa-value. Because fine

sand content Is found an Important factor, It was Included for evaluating R.

7. CUECK OF BEARING CAPACITY OF REINFORCED

CONCRETE PIERS FOR LATERAL FORCE

Even for a significant earthquake, the

bridge has to avoid critical failure.

Reinforced concrete pier which Is the most

important member for preventing critical

failure of bridge has to have the required

ductility as well as the strength. The check of

bearing capacity for lateral load was

Introduced to avoid brittle failure of

reinforced concrete piers in Inelastic range.

Check of bearing capacity for lateral load Is advised for the reinforced concrete piers with

small concrete section.

Fig. 5 shows a flow-chart of the check of

bearing capacity of reinforced concrete piers

for lateral force. Main steps of this check are

the following :

1) Judgement of bearing capacity for lateral

force

Bearing capacity for lateral force shall

be checked as

P« > kr.«W (9) where

Po : bearing capacity of reinforced con- crete pier for lateral force (tf)

khe : equivalent horizontal seismic coefficient for check of bearing

capacity for lateral load

W : equivalent dead weight (tf), and

shall be determined as W = Wu + CpWp (10)

).5 Pu^Pa

,1.0 Pu>P3

Wu : dead weight of a part of superstructure supported by the rein

forced concrete pier (tf)

Wp : dead weight of the reinforced con- crete pier (tf)

Pu : bearing capacity of reinforced con- crete pier for flexural failure (tf) Po : bearing capacity of reinforced con- crete pier for shear failure (tf)

When the check for the bearing capacity

for lateral force by Eq.(9) cannot be

Cp (11)

satisfied, modification of the design has to be

made so that the modified section, which is designed in accordance with tlie seismic

coefficient method, satisfies Eq.(9). It is generally advised to increase tlie concrete

section with decreasing the amount of

reinforcement.

2)Equivalent horizontal seismic coefficient

for check of bearing capacity for lateral

force

Equivalent horizontal seismic coefficient

for check of bearing capacity for lateral

force shall be determined as

kh<

k,

(12)

(13)

vT^Tn

kho = Cz • C| • Cr • knoO

where

khe : equivalent horizontal seismic coefficient for check of bearing ca- pacity for lateral force

khc : seismic horizontal coefficient for

check of bearing capacity for lateral force

Ij. : allowable ductility factor

Cz : modification factor for zone (refer

to Fig. 2) Ci : modification factor for importance

(refer to Table G) Cr : modification factor for structural

response (refer to Table 9) Knco : standard horizontal seismic coefficient for check of bearing capacity for lateral force

The standard horizontal seismic

coefficient knoa was determined so that it can represent a realistic ground motion developed

during a significant earthquake with

magnitude as large as 8.

3)Bearlng capacity for lateral force and

allowable ductility factor

Bearing capacity for lateral force Po and

the allowable ductility factor u shall be

determined based on the failure mode as : a)Flexurai failure

Po = Pv +

u = I *

-

Pu - Pv

a

<5 u - (5 V

(14)

(15) a S V where

Pu, 5 u : bearing capacity (tf) and ultimate displacement (m) for flexural failure

Py. (5 V : yielding force (tf) an yielding

257
dlsplament (m) for flexural

failure

a : safety factor (=1.5)

b)Shear failure

Pa = P. (16) U = 1 (17)

where

Pa : bearing capacity (tf) for shear fail- ure

8. DYNAMIC RESPONSE ANALYSIS

As number of bridges with complex dynamic

response characteristic increases, unified

stipulations for dynamic response analysis

become required. Although only general

consideration for selecting the Input ground

motion was described In the previous

specifications, a simplified but practical

input ground motion as well as an analytical

model and evaluation of the results of the

dynamic response analysis was introduced in the new specifications.

In principle, dynamic response analysis

shall be made by means of response spectral

analysis with use of an analytical model which

simulates dynamic characteristics of the

bridge. Acceleration response spectrum for

the response spectrum analysis shall be

determined as

S = Cz • c, • Cd • So (18)

where

S : response spectrum for response

spectrum analysis (gal)

Cz : modification factor for zone (refer

to Fig. 2) Ci : modification factor for importance

(refer to Table 6) Cd : modification factor for damping, and

shall be determined based on modal

damping ratio hi as Cd=— + (19)

40hi + 1 Sq : standard response spectrum for re- sponse analysis method (gal) (refer

to Table 10)

When the time history analysis is required,

strong motion records which have the similar

cliaracterlstlcs with S by Eq.(18) shall be used

with the consideration on site condition and

structural response of the bridge. Three

ground acceleration records which were

modified so that their response

characteristics match with So In Eq.(18) are

provided in Appendix 111. They can be used as the standard ground motions for the time

history analysis.

Because the results of the dynamic

response analysis highly depend on the

analytical model and the properties assumed

in the analysis, indiscriminate evaluation of

the results is quite difficult. Therefore it is required to use the overall result of tiie analysis considering accuracy of tlie properties and the analytical model used in the analysis. It is stipulated in the new

specifications that the results obtained by

the dynamic response analysis shall be used

for the check of seismic safety of the bridge

based on the allowable stress and

displacement which are referred in tlie seismic

coefficient method.

9. CONCLUDING REMARKS

The proceeding pages presented tlie outline of the main points of the new Seismic

Design Specifications for Highway Bridges

issued in February 1990. In tlie new

specifications, seismic coefficient method by

combining the previous seismic coefficient

method and the modified seismic coefficient

method is adopted, and based on tiie allowable

design approach this method Is regarded as the basic seismic design method. Besides the

design by means of the seismic coefficient

method, for those bridges which require the

checlc of ductility, the checlt of bearing

capacity of reinforced concrete piers for

lateral force Is made. Dynamic response

analysis can be made for the bridges which

require precise analysis considering

structural response.

REFERENCES

1) Kawakaml.K., Kuribayashi, E. , Iwasaki.T. and

Iida,Y.: On Specifications for Earthquake Re- sistant Design of Highway Bridges, 7tli Joint

Meeting, Panel on Wind and Seismic Effects,

UJNR. Tokyo, Japan, 1975

2) Kuribayashi, E., lwasaki,T., and Ueda.O.: New

Specifications for Earthquake Resistant De- sign of Highway Bridges In Japan, 12th Joint

Meeting, Panel on Wind and Seismic Effects,

UJNR, Washington. D.C.. U.S.A., 1980

258
Table 1 History of Design Loads for Hi(?hway Bridges in Japan

Design l ive Loads

Seismic Loads

k: Horizontal Scissmic Coefficient Year Name of Regulations Class

Truck Roller Loads

S tiee tear

Uniform

Loads

Line Loads

I m pact Loads Earthquake

1) 1886 Order No. 13 Ministry of Home

Affairs (MllA)

U^54 kg/m not

consi- dered

not considered

2) Order MHA

R-136 1

S-12 7

U-400 600 kg/in (carnage way) U-270 400 kg/rii (footway)

not con- derednot considered

3) 1919 Road Laws MllA R136 1 T-11.2 t S-30 t

Same as 2) - not

consi- dered

not considered

4) 1926 Specifications for Design of Roads. Road

Laws, MHA

1st

2nd

3id

T-12 I T 8 T T 6 I

U 600 kg/m

U-500 kg/m

U-500 kg/m

-

consi- deredSeismic Coefficient Mcihod K=0.15-0 4 depending on location and ground condition (V203 advised in Toyo. Yokohama)

5) 1939 Specifications for Design of Steel Road

Bndgcs, MHA

1st

2nd

T-13 1

T-9 T

U-500 kg/m

U-100 kg/m

consi- deredSeismic Coefficient Mcihod

k^=0.2, k_=0.1 1946. Nankai (M-8 1)

19-18. Hukui (M = 7.3) 1952. Tokachi-oki (M = 8.2) 1964

Niigata (M = 7.5)

6) 1956

(and 1964) Revision of Specifications for Design of

Steel Road Bridgci. JRA

1st

2nd

T-20 t

T-14 1

L-20 (5tm)

L-14 (3.5 t/m)

consi- deredSeismic Coefficient Method

k=0. 1-0.35

depending on location and ground conditions

7) 1964

lo 1971 Specincaiions for Design of Substructure of Road Bridges. JRA

Same as 6) Same k as (6) Dcuiled calculation methods

8) 1971 Specifications for Earthfiualte-Rcsisiani Design of Highway

Bndgcs. JKA

Same as 6) Seibinic Coef. method k=0 1-0.24 (Rigid) Modified SCM.

K-005-0,3 (llcxible) 1978 Miyagikenoki (M = 7.4)

9) 1980

Modil'ied SCM

Pan V Seismic Design Specifications

for Highway Bridges IRA '

Same as 6) Seismic Coef. Mcihod

k-0,1-0.24 (Rigid)

k=^0 05-O.3 (Flexible) Eanhquakc Response

Analysis (Very

[•Icxibic Bndgcs)

10) 1990 Same as 9) Same as 6) Seismic Coefficient Mcihod

k = 0, 1-0.3

Bcjring Capacity of RC

Piers for L^tcrai Load Dynamic Response

1982 Urakawa-oki (M = 7.1)

1982

Nihon-kai- chubu(M^7.7)

Tabic 2 Seismic Damage in l^cccnt Earthquakes

Miyor Eartluiuake Cluilij^c of M.ijor .Seismic I l.ilii.ig); l^'eisMiic Design Mellind Seisiiiie Inspect inn

iiiul Relri<niliii);

lOS.l Kaiito Carthcjuake (M7.9)

lO lG Nankai Eurtliriuakc (M8,I) 1948 Fukui Earlli.|iiake (M7.U)

1952 Tbkachl-okl Earthquake

(M8.1)

1064 NiigaU Eaithqualie (M7.5)

1978 Miyngi-ken-oki Earthquake

(M7.4)

1982 IJrakawa-okl Earthquake

(M7.1)

198:i Nihuii kai chulm

Earthquake (M7.7)

Hiilure of SiipirrNlrurliire due In Tiltiiig.'Moveiuent <if I'oiiiidatttiii

\iiliiic of Concrete around

"ixetl Mcaring

rijiiuige due lo

l,ii]iierjtttiin

Paihirc of UC I'iei^, and

liea^riny

I 1

H)lifi Iiiili.uidj) of Seismic Design (Sli|mlal inns for

Kn.iif ('i)ii^li ucl ion)

iO'.i'J lMlr(i(|ii(.li'jn of Sland.ud Seismic Cot'IficieiU

(SpucificQLloiis fur SU-cl liiidgc)

195G Seismic Coefficient depending on Zone and

Ground Condition (Specincaiions far Sleef

Bridge)

' Seismic Coefficient ilepcndiiig on Zone, Ground

Conditions. Importance and Slrtictural Kesponse

' Introduclioii of Kvalualion Method for Liquefaclion (Specifications for Seismic Design)

' i'iut V Seismic Design, Specifications for Design

of Highway Bi idge

hitrodiiction of New [Cvahiatlon Method for

Li<|iiefacli<tns

1000 l\irt V Seismic Design, Specificalins for Design

of Itoad Ih idge

1071 Seismic

hispeclion

107G Seismic

hispet lioti 1070 Seismic

Inspection

lOHG Seismic

Inspection

259
6 S U Q.

o

^.2

o

a 5

3 £- s 5

o

"S

it o

5-u

o o o o o o

•3 =>

11

5 ^

O O

O O

8 „

O o o o

o

o o

o o

a ~

a. c

- 3 o —

O. o

a. o < ^

o o o o o

o o o o o

o o o o o o

- " £-

C/1 S CO o o o o o o o

s g

c .5 o ^

;3 s Ido

J

2 i-

.5

il

3 u H a q 5 c

260
Table 4 Classification of Ground

Conditions

1980 Specifications 1990 Specifications

1 To < 0.2 (Rock) I To < 02

2 02 < To < 0.4 (Diluvium)

II 0.2 g T(. < 0.6 3 0.4 S < 0.6 (Alluvium)

4 0.6 < (Soft Alluvium) III 0.6 ^ T„

Tabic G Importance Factor C|

Group Definition

1st class 1.0

Bridges on expressway (limited access highways),

general national road and principal prefcctural road. Important bndges on genera] prcfectural road and municipal road.

2nd class 0.8 Other than the above

Table 5 Ground Condition Factor Cg Table 7 Structual Response Factor Ct

Ground Group I II III

Co 0.8 1.0 1.2

Ground Group Structural Response Coefficient

Group I T < 0.1

c.^ = 2.697'" > 1.00

0.1 g T g 1.1 c^ = 1.251.1 < T

= 1.33T-'"

Group II T < 0.2

c^ = 2.15T"'^1.00

0.2 ST S 1.3 c, = 1.25

1.3 < T

c^ = 1.49T-^

Group III T < 0.34

c^ = 1.80T"'>1.00

0.34 ST < 1.5 = 1-251.5 < T

a, = 1.64T-^

Table 8 Seismic Design Structual Unit

Lon^iliidtnal Tr;iiisversc Seismic Design Unit

Simple

(iirder

Coiilimioiis

Gir.ler

Ari'h, rrame anil Ollicrs

M F

A P;irt of Siipn-bliviclure wliiirli Causes Ucaclit)n F*)ri c

A r.irL nf Sup(.Tblru(ttiie which

Itcgardcd as A Unit Consisling of A

Siilistnicture and A I'ai t of Stiporbtniclnrc Siippoi tod by tlio Subs(riK.turc

T 1.)

nt'gardfd as A Unit

('otislsling of Subblnu tnrcs and the SiipcTstruftiire Supported by the

Sidjstniclurcs

261
262
LEGEND

: A c,= 1.0

]2] : B c, = 0.85

: C c, = 0.7

aJ.

^-0

?< US' 128' 130'

Fig.2 Seismic Zoning Map and Modification Coefficient Cz

2 g

0.5

0.3

0.2

0.1

0.07 O-OS"—

0.1

1 1 1 1—I—I I I I I

1 1 1—

r

•1980 Specs. (Group 1) 1990 Specs. (Group 1)

-I I I I I I I I I L. 0.2 0.3 0.5 0.7 1 Natural Period (sec)

(a) Group I

CO ^

N !—

0.5

0.3

0.2

0.1 c o = ^ 0.07

0.05

~T 1 1 1 I 1 1 I I

1980 Specs. (Group 2) 1980 Specs. (Group 3) 1990 Specs. (Group II)

_i I I I I I I I 0.1 0.2 0.3 0.5 0.7 1 2 Natural Period (sec)

(b) Group II

3 5

0.5

0.3

0.2

0.1 O o K O 0.07

0.05

0.1

-1 1 1 1—I—I

I I -1 1 r

1980 Specs. (Group 4) 1990 Specs. (Group III)

I I I I L. 0.2 0.3 0.5 0.7 I Natural Period (sec)

(c) Group III

Fig^.3 Comparison of Seismic Coefficient beti(?een 1980 and 1090 Specifications

263
TM -

to

T3aua

.9

OS o

l-l 8mouaun o

p—I

rtoo

D2 CL, O

0) ^ i- U O o ca J3 o o u

in

264
Scatter on Mechanical Properties of Structural Steels Recently

Supplied in Japan

Hirofumi AOKF

By

Hiroyuki YAMANOUCHF Ben KATO'

[Summary]

Statistical investigations are carried

out on mechanical properties of structural

steels and the causes of the scatter are

presumed. Higher yield ratios of steel

give unfavourable influences on failure

modes at the ultimate limit states of steel

structures, as well as the scatter of yield

strengths. The present paper emphasizes

that upper limits of not only yield

strengths but also yield ratios should be

specified in the appropriate industrial

standards in order to reduce the deviations.

Furthermore, it is necessary that accurate

measurement and standardized expression on stress-strain curves of steels. Such

statistical data are expected to be

accumulated internationally.

[Key words]

as nondeterministic. Therefore, statistical

data with respect to various factors are important to estimate the safty of

structures.

As uncertainty of the resistance, three

factors are generally considered, that is,

material, fabrication and professional [3].

The present paper concentrates a focus on randomness at the main stuctural material

of steel structures. First of all,

statistical features on mechanical

properties of structural steels recently

produced in Japan are surveyed. Secondly,

it is recomfirmed that plastic deformation

capacity of structural elements which

determines the ultimate limit state of

steel structures is strongly dependent on yield ratios of steels. Finally, standing

a view of point from structural engineers,

some reversions for the industrial

standards of materials are proposed.

Structural steel, yield strength, yield ratio.

Scatter of mechanical properties, ultimate

limit states, stress-strain relations

2. Statistics with respect to mechanical

properties of steels recently supplied

in Japan

1. Introduction

It is a worldwide prevalency that

structural design criteria are transferring

from allowable stress design to limit state

design based on reliable engineering.

Subsequently to Canada, Europe and the U.S.

[1], ^Standard for Limit State Design of

Steel Structures (draft) ' [2] was published

in this February by Architectural Institute

of Japan. In such a limit state

design, load and resistance are treated

The statistics on mechanical properties

of steels in the United States have been

investigated by Garambos et al[4]. The

statistical features on yield strength

(This is a term for yield point and off-set

yield stress.) and yield ratio which is defined as yield strength by tensile

strength were recently clarified in Japan

by analyzing the data by collecting mill

*^ Professor, Yokohama National University

*^ Head, Building Research Institute

*^ Professor, Toyo University

265
sheets which are issued as inspection

certificates from mill makers with blust

furnaces[5~7 1

.

Fig.-l indicates the distributions of

yield strength, tensile strength and yield

ratio of plates SS41 with less than 40mm in

thickness (24 kgf/mm^ is the minimum yield

strength and 41—52 kgf/mm-' is the range of

the tensile strength ) which are most used

as general structural steels. Referring to this figure, the mean value of the yield

strength is around 31 kgf/mm", which is

about 1.15 times of the specified minimum

value. The coefficient of variation is as much as 3.5%.

Fig. -2 also shows the distributions of

SM50 (33 kgf/mm' is the minimum yield

strength and 50—62 kgf/mm" is the range of

the tensile strength ) which is the most

popular high tensile strength steel adopted

for steel building structures in Japan.

Fig. -3 demonstrates those of steel

tubes with a circular or rectangular hollow

section. The yield strength of such a tube

rises higher caused by as cold-formed. As the result, the yield ratio also becomes

higher.

Furthermore, by analysis for these data

in detail, it becomes clear that yield

strengths are correlated with the thickness

of plates, as shown in Fig. -4. And, Fig. -5

shows the difference of the distributions

due to makers. Therfore, the distributions

shown in Figs.-1~3 can be variable if the

composion of the thickness of the plates or production ratio of the makers varies.

Above results can be summarized as follows.

1) Yield strengths of the plates with

thinner thickness is higher than that

with thicker one.

2) The distribution of tensile strengths is

normal, on the other hand, that of yield

strengths is approximately logarithmic

normal.

3) Scatter of yield strengths is large in

contrast with that of tensile strengths.

4) There exists a weak but positive cor- relation between yield strengths and

tensile strengths. Yield ratios are strongly correlated with yield strengths,

however, there is no relation with

tensile strengths.

The finishing temperature at the final

rolling for thinner plates is inclined

to be lower. This makes the yield

strength dependent on the thickness of

plates. Above 2) and 3) result from

that the upper limit of yield strengths

are not specified in the industrial

standard of materials.

3. Yield ratios of steel and Plastic

Deforoiation Capacity of Steel Elemewnts

It is indispensable to safty of

structures that the absolute values of

yield strength and tensile strength are secured as sufficient values. On the other

hand, scatter of the full plastic moments

of members is caused by scatter of yield

strengths. Therefore, it is pointed out that if the scatter of yield strengths is larger, the real collapse mechanism of the

frame becomes different from one which is predicted at the design analysis[6].

However, the yield ratio of structural

steels is a more essential factor for

ultimate limit states of steel structures.

That is, it strongly influences on the

plastic deformation capacity of structural

elements[7]. If the yield ratio is neally

equal to 1.0, a premature failure may occur

at the statically indeterminate structure.

Because redistribution of the moments is impossible owing to extremely small

rotation capacity at the plastic hinges.

Fig. -6 demonstrates the ultimate limit

states of cantilevers by comparing two different materials which yield ratios are

0.5 and 0.8, where buckling phenomena are

266
not considered. By observing this figure,

it can be easily found out that areas of

the plastic zones are different in spite of

the fact that the maximum moments at the

ends of the members are same. As a result,

the plastic rotation capacity of the member

with higher yield ratio becomes smaller. A statically indeterminate structure composed

by such members with higher yield ratio may

fracture before the collapse mechanism

completes, as shown in Fig-7.

Observing the statistical features of

yield ratios at Figs.-1~3 , the range of

yield ratios for SS41 is 0.55~0.85 and for

SM50 is 0.60~0.90. That for steel tubes

is especially large, because of effects of

cold-forming.

As clarified at the previous section,

yield ratios are strongly correlated with

yield strengths. Therefore, controlling

the scatter of yield strengths smaller is

most efficient for making the scatter of

yield ratios smaller. In other words, upper

limits of yield strengths and yield ratios

should be specified in the industrial

standards of materials. Table-1 lists re- presentative specifications for mechanical

properties of structural steels. Unfortunately in this table we cannot find any

standard which specifys an upper limit

of yield strengths or that of yield ratios.

Under international negotiations, these

specifications are expected to be realized.

4. Standardized stress-strain relations

A stress-strain relation of mild steel

under uniaxial tension loading has a clear

upper yield point as shown in Fig. -8. The

yield ratio can be definitely defined by

the ratio of the yield point to the tensile

strength. However, yield ratio of over 60

kgf/mm^ high tensile strength steel is

ambiguous, because the srtress-strain

curves of this kind steel is so-called

round house type. That is, off-set strength

is often used as the yield strength. It has been pointed out that plastic deformation capacity cannot be predicted properly

by such ambiguous yield ratios based on off

set strengths, otherwise that sufficient

complimentary energy is required for such

steels [9]. By other expression, the shape

of the stress-strain curve is also very

important factor as well as yield strength

and tensile strength. Therefore, it is insisted that stress-strain curves under

uniaxial tension loading should be measured

accurately up to the maximum load. With

these points as background, a tension

testing, specified standard shapes of

specimens and strain rates, has been

proposed by RILEM TC/83-CUS[ll] . According

to this method, stress-strain relations of

SS41 and SM50 are tried to be standardized.

Fig. -9 and Fig. -10 are the results and

Table-2 lists the characteristic values

which determine the standard curves [12].

5. Conclusive remarks

Statistical features on mechanical

properties of steels are investigated,

especially with respect to not only yield

strengths but also yield ratios which are most important factors for ultimate limit

states of steel structures. The results

are summarized as follows.

1. Concerning with yield strengths,

a) The scatter of yield strengths is due to the differences of production instruments,

technical level and quality control at mill makers.

b) Yield strengths are dependent on the

thickness of plates.

c) Yield strengths of the marerial as coldformed indicate very high compared with

the values specified in the design codes.

2. Concerning with yield ratios,

267
a) The scatter of yield ratios are larger

than that of yield strengths.

b) However, it is mainly caused by the

scatter of yield strengths.

c) Cold-forined members have very high yield

ratios.

d) The shape of the stress-strain curves or complimentary energy rather than yield

ratio is an important factor for steels

without a clear upper yield point.

Considering above conclusions, following

points are proposed.

1. The upper limit of yield strengths should

be specified in the industrial standards

of structural steels.

2. The upper limit of yield ratios also

should be specified.

3. Full stress-strain curves measured accu- rately up to the maximum tension load

should be accumulated.

[References]

[1] AISC :Load and Resistance Factor Design

Specification for Structural Steel

Buildings, Sep., 1986

[2] AIJ :Standard for Limit State Design of

Steel Structures (Draft), Feb., 1990

[31 Ravindra,M.K.and Galambos.T. V. :Load and

Resistance Factor Design for Steel, ASCE,

Vol. 104, No. ST9, Sep., 1978

[4] Galambos.T.V. and Ravindra,M.K. : Properties of Steel for Use in LRFD, ASCE, Vol.

104, ST9, Sep., 1978

[5] Aoki.H.and Murata.K. :Statistical Study

on Yield Point, Tensile Strength and

Yield Ratio of Structural Steel, Trans.

AIJ No. 335, 1984

[6] Aoki,H.and Masuda,M. :Statistical Investigation on Mechanical Properties of

Structural Steel Based on Coupon Tests,

Trans. AIJ No. 358, 1988

[7] Yamanouchi ,H. , Kato,B. and Aoki,H.:Statistical Features of Mechanical Properties of Current Japanese Steels,

Material and Structures, 24, 1990

[8] Kuwamura,H. and Sasaki, M. : Effect of

Randomness in Members' Yield Strengths

on the Failure Mechanism of Multi-Story

Rigid Frames, J. of Structural Engr. Vol.

34B,1988

[9] Kato, B. and Okumura, T. :Structural

Behaviour Including Hybrid Construction,

10th Congress of lABSE, Tokyo, Sep. , 1976

[10]Kuwaraura, H. and Kato, B. : Inelastic

Behaviour of High Strength Steel Members with Low Yield Ratio, Pacific

Structural Steel Conference, May, 1989

[lllRILEM TC/83-CUS : Tension Testing of

Metallic Structural Materials for

Determining Stress-strain Relations

under Monotonic and Uniaxial Tensile

Loading, Material and Structures, 23, 1990

[12]Kato, B., Aoki, H. and Yamanouchi, H.

: Standardized Mathematical Expression

for Stress-strain relations of Structural Steel under monotonic and Uniaxial Tension Loading, Materials and

Structures, 23, 1990

268
25-

15-

10

5H

X=45.73 5=1.89

SSAIVSMAl

Ti=5S06

^=68.26 n 5=5.67

Nomdl D'stribulion Logarilhmic Nomot

Dislribulion

25 • 30 35 ^0 i,5 50 55 60 -65 70 75 80 85 90

YPJS(I<gf/mm') YRCV.)

Fig. -.1 Histogram of Mechanical Properties for SS41 and

SM41 (6mm<t<40mni)

X=55.01 SM50

-tiAO YP:Y;e!d Strength

TS -tensile Strength

YR:Yield Ratio X -'Mean n=3773

25 30 35 ^0 45 50 55 60 65 70 75 80 85 90 95 100 YP.T5(!<gf/mm^) YR(V.)^

Frg;-:2 Histogram of Mechanical Properties for SM50

STKAI.STKRAt

n=596

X=81.59

5=6.28

25 30 35 40 45 50 55" 60 65 70 75 80 85 90 95 100

YP.TS(kgf/mm*) YR(Vj

Fig. -3 Histogram of Mechanical Properties fpr.STK41 and STKR41

269
f-l a

CS CS (S

OO PN CD

(X) Hk <

(3

CO

f-l

>-l a

c

«j

U2<UPi

c

wc

o

•1—1 M

cd

r-HOJ

I60

•i-H

270
o<Q

O

>»

-P

•HO

w

cd 0)

cd o •H c <D o o U

•H •H

-P +J P

cd

m

O

-p cd o rH

a; P-. 4J o

•H

P>

CdE

FhO

<M

<D Q

4J

cd

-p

•HO

cda,

cd o

CO

Cmo

>5

+J•HH

rT

-pb3 X

cdE CDE

cd

fH

•H

bJ3 •H

cc:oE

02 •H

cd |x; 1 o' Q)

''X.

X

cd <DW

Ph

CD rHOO

c—1 bO

00o

II

X

aJ 1 E

CL,

XE

2: ; X, cdE

''P-i\ o

II

Xe

CmO

-P

•H

CJ

cda.

cd oo

•HP

cdeo

0) Qo

•HP

cn

cd rH

Ph

COI oH

-PO

(D CO

cd

-p

rHm

CD>

rH

•H

-P

cd o

271
Fig. -8 Shapes of Stress-Strain Curves

272
(kgf/mm^) SS41

- m ^ — y

^ —

1—A'/

1

. J

• -

' - • m - 2 s-- -

a:Hean s:St andard deviation

0.1 0.2 0.3 "~ "1

Fig. -9 Standardized Stress-strain Relations for SS41

(kgf/mm^) SH5dA

I

• ra+2

1 s

s

1- - 1 ^^tV^^ \ —r i

/

. / _;

1

i:Mean s: Standard deviation

0. 1 0.2 0.3

Fig. -10 Standardized Stress-strain Relations for SM50

273
Table-1 List of Standards

Country Standards De s Ig—

nation

Mln. Yield

Stress

Tens

.

Strength Y* Remfl rks

MN/m^

0 - 245 412 -v- 490 0.594 INTERNATIONAL ISO

Fe 52 343 490 -x, 608 0.70

a Grade AO 230 400 •V 480 0.575 'ENGLAND . BS 4360

Grade 50 Grade 55 345 430

500 550

620

700 0.69 0.782 } Co lutnb turn flnd/oir V«infldl.uro

Grade E24 235 363 -x, A41 0.647

Grade E30 294 461 v- 559 0.637 FRANCE NFA 35-501 Grade E36 353 501 '^' 608 0.692

- Grade A50

Grade A60

Grade A70

294 333

363

490

588

686

'^' 588

706 833

0.60 0.567 0.530

Tmnnrr r^^rc Nnr Hpn'rl

Limited Weldabllltv

JIS ^SSAl ,SM41 235 403 510 0.583

C - 3101

-v, 3106 SH 50A,B 315 490 -v- 607 0. 642

- SM 50Y 353 490 '^- 607 0.72 Columbium and/or Vanadium

JAPAN SM 53 353 520 •v- 642 0.678

SM 58 452 570 '^ 718 0. 794 WES HW 70 686 785 •^. 932 0.874 HW 80 785 864 -^-1030 0. 908 HW 90 883 951 1130 0.93

"a 36 248 400 -v- 552 0.62

A242: weathering

A242, A4A1, 317 462 0.686 steel

AAAAl U.S.A. ASTM A572 Gr 45 310 414 0.748

50

55 345379 448

483

0.77 0.785 Columbium and/or Vanadium

60 414 517 0.8 A 588 345 483 0.713 Weathering Steel

A 5U 690 793 -N. 930 0.87

"st 37.283 240 370 450 0.647 WEST

GERMANY DIN 17100 St 46.223 290 440 540 0.658

St 52.3 360 520 620 0.691

* Y - Yield Ratio;not specified in these standards.

274
Table-2 Material Constants for Standardized Curves

GRADE OF

STEEL

kgf/mra^

£ 9 8 - £ m £93 + R ( 7 ) fyh fyl f max

SS41

in + 2S 34 32 45

0.14 0.235 0.33 0.50

(1.28)

m 29 27 43 (1. 27)

[n-2S 24 22 41 (1.26)

SM50A

m+2S 43 41 56

0.11 0.19 0.27 0.57

(1. 23)

m 37 35 53 (1.22)

in-2S 31 29 50 (1.21)

in: Mean value s: Standard deviation

0^£^£y cr=E-£. £y = f^f,/E

£ s--^ £ ^ £=' a= (1)

£n'^£^£n (7 =fmax

c =

c R 1/ R {!+(—) } £ 0 £0=7 'fmax/E

(1)

(2)

275
Development of Effective Use of New Metallic Materials

for Building Structures

by

Hiroyuki Yamanouchi^ Hirofumi Aoki

and

^ Tatsuo Murota

ABSTRACT

The Japanese Ministry of Construction has

initiated a five-year research and

development program on "Effective Utilization

of Advanced Materials for Construction" from

the fiscal year of 1988. As part of the

national program, new metallic materials for

building structures are also adopted. This

paper describes the outline of the metal

phase of the program and issues to be solved

in the future work.

KEYWORDS: New metallic materials; building

structures; technology of effective use.

1. INTRODUCTION

Recently in Japan, it is widely recognized

that buildings and facilities having high

quality and good performance should be

constructed to establish more durable and

usable social capital. As a necessary

consequence, this trend requires building

materials to be more durable,

fire-protective, mechanically excellent and

beautiful. These days, thus, new materials

having new function or performance have come

to be partly or fully used for buildings,

even if these materials have also higher

cost

.

The national program intends to support

technically the above recent trend and then

to lead to new technologies that enable new materials to be effectively used for

buildings

.

As part of the program, some new metallic

materials for building structures are targeted for developing new building

technologies. This paper describes the

outline of the metal phase of the program and

identifies technical issues to be solved in

the course of the program

2. PRELIMINARY STUDY

Before discussing the topic on each new metallic material, we should describe briefly

preliminary studies which were carried out to choose new materials to be targeted in the

future work of the program. As is well

known, there are all kinds of new materials

which have been developed in many industrial

fields. Thus, as the first step of the

program, it was strongly needed to find out

new metallic materials highly applicable to building structures; these materials should ' also be under a feasible outlook on developing technologies for effective use within the limited period of the program. In the first year of the program, the following

ten materials were then chosen for the

preliminary studies;

1) Structural steel with high strength and

low yield ratio (high quality steel)

2) Stainless steel

3) Aluminum alloy

4) Laminated high-damping steel plate

5) Ultra mild steel

6) Cables

7) Clad steel

8) Steel-concrete composite material

9) Steel-rubber laminated bearing for base

isolation

10) Titanium alloy

Regarding these materials, the present

condition of development or production,

material properties and applicability to building structures were studied through

hearing from specialists of production areas

and with extensive literatures.

3. TARGETED MATERIALS

On the basis of the preliminary studies, we finally determined the materials to be

adopted in the program. In the selection,

the following criteria were adopted; a material coming under at least one of the

criteria should be eliminated:

1) Design manuals or specifications are well

established

.

2) Administrative treatment is now necessary

rather than research and development.

3) The cost is too high for application to building structures.

A) The production system is not fully

completed

.

5) The material has low usability for major

structural members.

6) Production and user sides have less concern over the use of the material.

Some of the above items should be explained

in more detail. That is, materials

1 Building Research Institute, Structure

Department

2 Yokohama National University

276
applicable to the item 1 have good design

manuals or specifications on the basis of

basic research. Thus, there is almost no need of a further study in this program; this can be said even if those manuals or specifications were established by unofficial

groups. Also in the case of item 2,

necessary technical development has been

almost finished so that an administrative

treatment is now needed for general use in

building structures. Materials coming under

the item 3 should be eliminated from target

materials since building structures use a large amount of material in general. Thus

the use of such materials would extremely

boost the total cost of the buildings.

Regarding the item 4, more stable and wider

supply should be considered. For instance,

if only one maker can produce the material,

it is not appropriate to develop a utilization technology for the material in

the national program.

By applying the criteria to the ten materials

mentioned in the preceding section, the

following five materials were selected as target materials in this program:

1) Structural steel with high strength and

low yield ratio (high quality steel)

2) Stainless steel

3) Aluminum alloy

4) Laminated high-damping steel plate

5) Steel-rubber laminated bearing for base

isolation

Here, the reason for eliminating the rest

five materials should be briefly noted. That

is, the "ultra mild steel" which has the

lower yield strength around lOkg/mm^ and high

ductility can be supplied by only one steel

makers; this material corresponds to the

item 4 of the criteria. However, this

material would be usable as a material for

"damping devices". It is thus hoped uhat more steel makers produce this type of steel

for stable supply. The "cable" has already

good design manuals in the field of bridges

in Japan. In the field of building

structures, thus, an administrative

consideration is now needed to make the

material generally usable. The "clad steel"

has also well-established standards in the

field of chemical plants and machinery. In

the world of building structures, however,

the need for this material is not dug up as yet. Thus it is too early to pick up this

material as a target. The "steel-concrete

composite material" is in a somewhat

different category from the other materials

because this material has already been used

widely in building structures. However,

regarding some types of structural members

made of this material, for instance steel

tubular columns filled with concrete, there

exists a significant discrepancy in

evaluating member strength, between the

design recommendations of the Architectural

Institute of Japan and the Enforcement Orders

of Japanese Building Standard Law. However,

this different view does not need more research and development, but probably a careful administrative treatment for general

use. The "titanium alloy" is too much

expensive, about fifty times as costly as ordinary steel, whereas this material has

very excellent performance. Then, the

usability of this material may be very low

for general building structures. However,

this material would be useful for localized

use in structures under special environment

such as an offshore condition. In fact,

titanium alloy bolts have been developed for

the use under such severe conditions.

4. ISSUES TO BE SOLVED

In this section, we describe the present

proceeding of the program and issues to be

solved for each targeted material.

I) High quality steel

This steel has considerably high strength

(60kg/mm^ in maximum tensile strength) , but

Che yield ratio, defined as the ratio of

yield strength to maximum tensil'e strength,

is less than 0.8. Furthermore, the yield

strength is well controlled with the range of

48±3 - 5kg/mm^. Japanese major mill makers

have already been able to produce this type

of steel by the Thermo Mechanical Control

Process (TMCP) which is applied in the latter

stage of rolling.

By a trial-design study in the program, it is

found that 10 to 20 percent of the amount of

steel can be saved when this new steel is

used in lieu of normal steel with the yield

strength of 24kg/mm2. However, this saving

would not always ensure the spread of this

steel because the cost of the steel is

expected to increase by 10 percent or more. Furthermore, the cost of building structures

themselves, now in general, amounts to around

20 percent of the total building cost. On

the other hand, the use of the new steel can

give a considerable advantage to structural

engineers since the steel would provide

building structures having good performance.

However, this kind of merit cannot be easily

understood by building owners. Then, the key

points for effective use of the new steel

would be as follows;

1) Development of new structural systems by

using the new steel

2) Simplification of structural design and

shortening of construction terms

Regarding the first item, it is necessary to develop new design requirements which reflect

277
the peculiarity of the new steel. The second

item is related to the smaller scatter band

of the yield strength of the new steel than

that of normal steels. It is a key point

whether or not the well-controlled yield

strength would make design procedures simple

and also make inspection work easy.

2) Stainless steel

As is well known, stainless steel has not only higher corrosion resistance but also

larger ductility and lower heat conductivity,

than normal carbon steels. On the other

hand, a weak point of this material for

general use is its higher cost; at present,

four to five times as expensive as normal

structural steels. Probably for this reason, stainless steel is not prevailing so far for

structural members.

Stainless steel has a lot of kinds

corresponding to the combination of chemical

components. Then, the two thirds of all

product in Japan is the JIS grade of SUS304

(I8Cr-8Ni) which is austenetic stainless

steel and may be most suitable for structural

members. In other grades, the grade of

SUS304N2, which includes nitrogen, and the

grade of SUS329J2L, which is

austenetic-f erritic duplex phase stainless

steel, are also applicable to structural

members as high strength stainless steels,

whereas these two grades have some problems

in welding and forming.

The first problem in making use of stainless

steel for structural members is whether or not stainless steel members can be used

without fire resisting covering; designers

would desire to make the best use of the fine

surface of stainless steel. For this, first,

thermal characteristics of stainless steel

members should be carefully investigated.

Second, the Japanese method for fire

resisting design should be revised so that

the thickness of fire resistive covering can

be determined corresponding to the thermal

stress and the surface temperature of

members. Yet, since this is a drastic change

in the fire protective and resistant design

philosophy of Japan, it will take a long

period over the remaining years of this

program.

The second key is directly related to the cost of stainless steel. The present cost is even higher than ordinary structural steel as mentioned before. According to the

preliminary study, there is little prospect

of being able to cut the cost by a large

margin even after ten years. Thus, it may

not be reasonable to build a structure fully

with stainless steel. Otherwise, stainless

steel may be used partially in structural

members to make use of its high ductility;

for instance, in ordinary steel members,

portions at which hinge formation is expected

may be replaced with stainless steel to gain

more hinge rotation.

The third point at issue is in the present

fabricating and constructing system of

stainless steel structures in Japan. At present a few fabricators can deal with

stainless steel to make up building-structural members. Even these

fabricators have not always good quality

control system in their factories.

Furthermore, a qualifying match system of

welding workers is not established as yet.

Now, in order to diffuse stainless steel as structural members, the establishment of a fabrication and construction system

sufficiently assured is urgent business of

concerned areas.

3. Aluminum alloy

Japan now imports about 90 percent of all

aluminum alloy ingot which is processed in

Japan. From thirty years ago, the Japan

Light Metal Association has dealt with how to make good use of aluminum alloy for

structural members. On the other hand,

before the above activities, the Japanese

Architectural Institute published "Design

Guidelines for Aluminum Alloy Structures" as

a draft version. Unfortunately, however,

these activities did not effectively spread

the material in the field of building

structures, in particular in the use for

major structural members; aluminum alloy is not yet approved as a material for major

structural members in the Japanese Building

Standard Law. As recent examples of aluminum

structures in Japan, we can cite a covering

frame over a heated pool, top-light frame,

greenhouse frame and sunroom frame;

nevertheless major structural members are made of steel in most cases. The keys for

developing utilization technologies on aluminum alloy are similar to those on stainless steel. Here the problem of fire

protection and resistance for aluminum alloy

is more serious than for stainless steel,

since the melting point of aluminum alloy is even lower than that of stainless steel;

that is, around 620°C.

On the other hand, regarding the fabrication

and construction of aluminum structures,

aluminum extruding companies have

sufficiently covered this area so that there

exists almost no problem such as poor quality

of work. As a technical point, finally, an advantage that free sectional shapes can be

obtained by extruding should be effectively

used. For instance, a wall structure

comprised of a thin walled wall-like box

section including inside stiff ener plates,

may be a structural system pertinent to

278
aluminum alloy.

4. Laminated damping steel plate

This damping plate is comprised of two thin

steel plates and a thin layer of viscoelastic

polymer that is adhesively sandwiched between

the plates. When the damping plate vibrates,

vibrational energy is transformed into

thermal energy by a shear deformation in the

polymer, resulting in the vibration highly

damped. In particular, the damping plate

seems to be most effective in preventing a solid-propagating noise. Stainless steel is

also used for the thin steel plates.

In Japan, major mill makers are producing the

damping plate for the use of floor plates,

roofing plates and exterior stairs. As the

present problems in fabricating the damping

plate, there is some difficulty in welding

owing to combusion of polymer and in press

forming owing to spring-back which disturbs,

in particular, right-angle forming at corners. Then, the keys for effective and

wide use are to determine a common testing

method for evaluating damping

characteristics, and to standardize a method

for evaluating the performance of each used

portion in building.

5) Laminated steel-rubber bearing

Laminated rubber bearing is made of thin

steel plates and rubber sheets by pasting one over the other alternatively. In Japan,

major rubber-tire makers have initiated to produce this type of bearing for the use of

seismically base-isolated buildings. By

using the rubber bearings, natural vibration

period of a building can be made extremely

longer since the bearing has a low horizontal

(shear) stiffness; the bearings support the

building weight with high vertical stiffness

owing to the stee''. plates in the bearing.

The types of the bearing are classified as follows

:

1. Ordinary type mentioned above

2. Lead-rubber bearing

3. High damping rubber bearing

The first type bearings have low inherent

energy absorption so that additional dampers

are usually needed to form a base-isolated

structure. The lead-rubber bearing contain a lead plug inserted in the center hole of the

bearing as an incorporated damper. The

rubber itself of the third type bearings has

high inherent damping capacity enhanced by

mixing filler.

In Japan, most research and development are being made to summarize design guidelines or

to accomplish safety assessment methods in

the following organizations:

1 . Tri-partite R&D among the Housing Bureau

of the Ministry of construction, the

Building Research Institute and the

Building Center of Japan; to develop a safety assessment method for technical

approvement of base-isolated buildings

2. The Architectural Institute of Japan; to summarize design recommendations for

base-isolated buildings

3. The Public Works Research Institute; to make progress in the technology of

base-isolated highway bridges.

On the basis of the above progress, the

significant keys in our new program in then

to standardize the testing method for

characterizing rubber bearings, to grasp

the scatter of performance characteristics

of products, and to evaluate aging

characteristics and degradation of rubber.

In these activities, it is hoped to reflect

requirements of structural engineers as the user side.

5. SUMMARY

The program status and future work are briefly introduced concerning each new metallic material for building structures.

As initial conclusions on the basis of

preliminary studies, the following three keys

are found to be significant commonly for all

materials

1. Development of new structural systems

using new materials

2. Cost implications

3. Control and assurance of quality of

products, and fabrication and

construction work

The program is now in the beginning of the

third year so that many sorts of experimental

studies are in progress to draw design

recommendations on structural members with

new concerned materials. After three years,

then, some of those materials will be

recognized as "existing" materials not only

by structural engineers but by the building

administration.

279
I
Theme V

Loma Prieta Earthquake

Initial Overview of the San Francisco Bay and

Santa Cruz Mountains Ground Motion

by

A. Gerald Brady

ABSTRACT

The strong-motion accelerograms from the Loma

Prieta earthquake are analyzed for their longperiod content in order to obtain a clearer

picture of the long-period wave propogation

details. Shear waves having periods in the 3.5

to 4 sec, and 5 to 7 sec ranges travel across

four groups of stations with satisfactory

coherency. Displacement accuracies are of the order of 0.5 cm for most of this data, with

signal amplitudes an order of magnitude higher

than the noise. Resonances associated with

shear waves of 1.5 sec period are responsible

for about 3/^ of the differential displacement

necessary to unseat the 15 m section of the Bay

Bridge

.

KEYWORDS: Bay Bridge; coherency, displacements;

earthquake; shear waves.

1 . INTRODUCTION

Strong motion accelerographs from at least 139 stations were triggerea by the MgT.l main shock

of the Loma Prieta, California, earthquake

occurring at 00014:15.24 GMT, on 18 October

1989^1 •2.3, _ The hypocentral depth was l8.5

km, deeper than most events on the San Andreas

in this region. During the 7 to 10 seconds

following initiation, the rupture spread

bilaterally along the fault, and upwards towards

the ground surface on the 70° downward-dipping

fault. The total ruptured section is estimated

from aftershock studies to be 40 km in the northwest-southeast direction, equispaced about

the hypocenter , and 13 km deep, failing to reach

the ground surface by about 5.5 km. In the hypocentral region the relative motion along the fault is estimated to be 2 m horizontally and

1.2 m vertically; the southwestern side moved

northwestward and upward with respect to the other side. The right-lateral displacement is typical of San Andreas movement; the vertical

displacements are not, although the non-vertical

dip and the slight bend in the San Andreas

alignment in this region will have contributed

to this effect. The vertical displacements are certainly coordinated with the building of the Santa Cruz Mountains, in which the ruptured

segment lies.

Although the 139 stations with triggered

accelerographs are scattered throughout the San Francisco Bay area and areas to the east and

south of the epicenter, some specific groups of stations afford the opportunity to study the transmission of long-period waves along the ground surface, or upwards from the rock

below. This study is oased on the appearance of waves with displacement amplitudes of several

tens of centimeters and periods of several

seconds during the routine processing search for

a long-period (low-frequency, low-cut) filter

that leaves relatively untouched those frequency

components whose signal levels are well aoove

noise levels.

Part of this search for a long-period filter

relies on the fact that over an array of stations of small enough geographical dimension,

waves of sufficient wavelength (period times wave velocity) must appear coherent as they cross the array, that is, they must retain

closely their shape and amplitudes, although

they may be offset in arrival times by small amounts

.

Another of the more important criteria used in determining the long-period filter is a comparison between the Fourier amplitude

spectrum (FAS) of a signal including its noise

and the spectrum of a representative noise

sample. At the long period where these spectra

merge, noise is clearly prominent and a lower

period must be chosen for a final processing

filter

.

Processing of the Loma Prieta strong-motion data

continues at both the California Division of Mines and Geology's Strong-Motion

Instrumentation Program (CSMIP)^-* at Sacramento, and the U.S. Geological Survey's

Cooperative Network of Strong-Motion Instruments

(USGS)^ ' at Menlo Park. This investigation

studies the processed data from arrays shown in Figure 1 , at

:

1.1 Anderson Dam, 27 km from the epicenter, on a line perpendicular to the San Andreas through

the epicenter, and covering several kilometers;

1.2 Gilroy Array, Numbers 1, 2, 3, 4, 6, and 7, centered approximately 31 i<m east of the epicenter, but spanning from 15 to 28 km from

the nearest part of the rupture, measured along

a line approximately straight and transverse to the San Andreas.

1.3 Stanford, Palo Alto, and Menlo Park. A triangular array 51 km in a northwesterly

direction from the epicenter, and centered 7.3

km from the closest point of the San Andreas,

although the rupture did not reach this point.

The triangular array, with one station at its center, has sides between 6 and 7 km long.

U.S. Geological Survey, Menlo Park, California,

94025 USA

283
1. '4 San Francisco to Oakland. This array spans

15 km from Che Franciscan rock in San Francisco

to the Recent Alluvium in Oakland, roughly

following the line of the Bay Bridge. This line

is approximately 97 km from the epicenter, and

is aligned transverse to the radial direction.

Its center lies 20 km from the nearest point of the San Andreas.

All processed data from these stations are available from CSMIP (CDMG, 63O Bercut Dr.,

Sacramento, California) and the USGS (Menlo

Park, California). Table 1 contains the details

of the stations and instruments in the arrays,

with their routinely processed long-period

limit

.

2. ACCURACY OF DISPLACEMENTS

The accuracy of displacements calculated by

integration from digitized accelerograms has been the subject of many investigative efforts, ever since routine processing of accelerograms

was performed at Caltech in the early

1970s^^ . A description of the accuracy of the long-period amplitudes is required for a study

of the displacement histories at more than one station as waves traverse a specific array. These displacements are characterized by longperiod content. Consequently long-period errors

are of interest. The discussion is confined to errors derived from the broadband white noise errors in acceleration during the digitizing

process. The causes of these errors include

accuracy In the selection of the centerline of

the trace and round-off error associated with

the least count in the Y-direction^ .

Sources of error in the digitized acceleration

trace that are removed by subtraction of a digitized reference trace include inaccurate or non-rectilinear film transport during recording,

distortion during film processing, and

inaccurate but consistent and stable digitizing

hardware. These hardware problems include

transportation of the digitizing mechanism

relative to the digitized film, and optics

problems including parallax. The reference

trace to be subtracted is assumed to be placed

close to the data trace. These errors are not discussed further.

The CSMIP data discussed here is produced on a rotating drum raster scanner whose optimum

operational characteristics incorporate a pixel

size in the Y-dlrection of 50 micron. A film

trace sensitivity of 1 .9 cm/g results in a least

count of 2.6 cm/sec . Descriptions of tests on this system indicate that the RMS error is of

the order of one count, or 2.6 cm/sec . The uses data reported on here is produced on a trace-following laser scanner, whose least count

in both directions is one micron, corresponding

to 0.052 cm/sec . Digitization of very small

amplitude records^^^ and of typical reference

traces confirms that the RMS error, depending to some extent on the quality of the focusing of

the traces, varies between 5 and 15 micron. The

great majority of traces have very high quality,

with an RMS error of 5 micron, corresponding to 0.26 cm/sec .

These RMS errors in acceleration can be translated into RMS errors in displacement. The

FAS of a digitized reference trace, treated as acceleration, is relatively flat throughout the entire frequency range of this study, 25 Hz to

10 sec, including particularly 0.1 to C.25 Hz, and the displacement spectrum would consequently

fall off approximately with the inverse square

of the frequency. This results in the RMS displacement errors in Table 2 for the two digitizing procedures, at periods between H and

10 seconds. These errors are not necessarily

those in listed peak displacements; they

indicate the error in oscillatory disp! "cement amplitudes during oscillations at the listed

periods. These expected errors must be recognized in the following sections.

3. DISPLACEMENTS ACROSS SPECIFIC ARRAYS

3. 1 Anderson Dam

Processing the Anderson Dam records permits a full description of the decisions made on the long-period content. These are representative

of all the processing in this study. The

processed film records are from instruments at the crest, left abutment, and downstream of the earth and rockfili structure, 64 m high with a crest length of 422 m. The abutment and

downstream instruments are approximately 300 m and 250 m, respectively, from the crest instrument

.

An accurate measure of the long-period noise

effects can be obtained in the spectral domain

as follows. When a reference trace is subtracted from a data trace the result is effectively the sum of the noise-free signal and

a quantity equal to the difference between two random error histories. At long enough periods,

the noise will be dominant, and the FAS will so Indicate, with increasing amplitudes. A close

approximation to the spectral content of this

difference between two random error histories

can be obtained by subtracting the relevant

reference trace from a second digitized

reference trace, preferably closely placed on

the original record. At long enough periods,

this spectrum will therefore have the same increasing amplitudes as the previous one. The spectra meet and coalesce at periods where noise

is dominant over the signal. Figure 2 compares

the FAS of the left abutment 243" component

(with no filtering, but with reference trace subtraction and mean value subtraction) with the

FAS of another reference trace from the same

film, prepared in the same way. The spectra

have barely merged at 40 sec (0.025 Hz), a period equal to the record length. At '0 sec

(0.1 Hz) there is at least an order of magnitude

between these spectra.

Filtering at a series of long-period cut-offs.

284
namely, 4, 6, 10, and sec, showed that 10 sec waves crossing this 3"3tation array remained

coherent. This is shown in Figure 3, where the displacements at the three stations are offset

appropriately in the Y-direction for clarity and

in the X-direction for alignment in time -- these instruments are not wired for simultaneous

triggering or common timing marks. Comparison

of these displacements with the value for the uses RMS error of 10-second waves, namely 0.66

cm, is good. Another feature illustrated by

this figure is the shape of the corrected

displacement waves at the time of triggering.

The bi-directional Butterworth filter performs

with no obvious distortions and no ringing; the displacements from the non-simultaneous

triggered records line up well.

The shear waves arrive at Anderson Dam 3-5 sec after the first instruments triggered. Although

shear waves of all periods arrive at practically

the same time, the 1 O-sec filtered displacements

exaggerate the waves of longest period. The

3^0° component, along the dam crest and

approximately parallel to the San Andreas, shows

a probable shear wave of period 5 to 7 seconds,

having initial motion in a direction close to southeast. Radial motion, not clearly

associated with these shears, also has prominent

6 sec periods.

3.2 Gilroy Array

The long-period limit chosen for these CSMIP records is 6 sec, using a filter whose response

falls to zero in the interval between 6 and 12.5

sec. The north and east components of all six

stations indicate a sharp shear wave arrival in the northeast direction, 3 sec after triggering

at #1 , and 5 sec after triggering at #6. This

is probably a transverse shear wave, initial

motion in the NE direction, the wave front

moving down the fault in the SE direction. The

amplitude of the northeasterly swing in displacement reaches 22 cm in the alluvium

deposits in that part of the array closest to the San Andreas, and falls to a few centimeters

at the array's eastern end, 28 km from the fault.

This shear pulse provides the prominent longperiod ground motion to the southeast of the earthquake. The displacement plots indicate

that at each station, after the initial ground

motion to the northeast, a number of waves with

specific periods were prominent. With the help

of the FAS plots, these periods are found to be 1.4, 3 and 5 sec

.

3.3 Standard, Palo Alto and Menlo Park

Four stations forming a triangle with a central

point provide records that illustrate the ground

motion in the vicinity of Stanford University.

Triaxial sensors at ground level recorded the motion at a Stanford parking garage (SPG), the central point, 51.3 km in an approximate

northwesterly direction from the epicenter. The

stations at Stanford SLAC, the Menlo Park VA Hospital, and the Palo Alto VA Hospital are each

approximately M km from SPG. Three of the stations indicate that ground motion during the passage of a 3-5 sec wave was transversely

aligned with the epicenter; a small vertical

component to the motion is also present. The peak displacements at the four stations depend

on local geology effects, resonant amplifications and waves at shorter periods, but

the amplitude of this specific long-period wave

lies in the range of 12 cm at SLAC, closest to

the San Andreas, to 8 1/2 cm.

Coherence is more difficult to identify in the accelerations. Several occurrences of 1 Hz content is identifiable across all four stations

in their most east-west trending components.

Peak accelerations, however, are associated with

higher frequencies than this, affected also by

local geology, resonances, simultaneous arrivals

of peak amplitudes from higher frequency waves, interaction with the structure housing the instrumentation and other nearby structures, and other considerations.

3.4 San Francisco to Oakland

The longitudal motions of the eastern portion of the Bay Bridge between Yerba Buena Island (YBI) and Oakland are estimated from the main shock

records from accelerographs at YBI, the Oakland

Outer Harbor Wharf (OHW) and the Pacific Park

Plaza at Emeryville (EMV). YBI is on Mesozoic

sediments of the Franciscan formation and the recorded long-period motion may be considered

representative of the western abutment of the cantilevered section of the Bay Bridge and also

representative of the lowest levels of the concrete piers under the cantilevered section

and the steel truss section in contact with this same hard rock. OAK and EMV are on filled land overlying Recent Alluvium; their motion is representative of the eastern abutment of the

Bay Bridge in Oakland and the lowest levels of

the piers supporting the bridge near the abutment which are in contact with the same

sediments.

The records were filtered at a long-period limit

of 6 seconds, and rotated to an 80' aligninent.

Only OAK had radio time (trigger time was 0004:35.0 GMT on 18 October 1939) so that alignment in the X direction was made instead on

the arrival of the S wave, using the accelerogram traces and concentrating on 3 to 4 Hz content. At the same instant, the displacement plots show the arrival of a longperiod wave of 3-5 to 4 sec period, 5.8 cm amplitude, transversely aligned with the epicenter, that is, longitudinal to the bridge. The hard rock sites in San Francisco

(namely Pacific Heights and Rincon Hill), YBI, OHW and EMV all contain this ground movement

with remarkable coherency. A higher frequency

wave (period 1.5 sec) is also barely present,

riding on the 3-5 to 4 sec wave, on the rock

sites. At EMV and OHW this has resonated to

285
amplitudes of 8.6 and 12.7 cm, respectively,

still riding, now in an overpowering way, on the longer wave. If the assumption is made that OHW

is representative of the longitudinal abutment

motion, subtraction of YBI from OHW yields a differential displacement peaking at 11 cm between the two bridge abutm.ents (Figure 4). Caltrans measurements^ indicate that the ends

of the two trusses, where the 15 m bridge deck

slabs failed, separated by 26 cm and returned 13 cm. The bearing seat was only 15 cm wide. As reported here, measurements from the records

indicate that 11 cm of separation occurred at the foundation level. Structural dynamics

during the earthquake must account for the remaining Caltrans separation.

U. CONCLUSIONS

Although ground accelerations recovered from

strong motion accelerograms provide a means for calculating the forces suffered by the structures affected by the Loma Prieta

earthquake, a better understanding of the overall ground motions emanating from the epicentral area has been obtained by considering

the long period content portrayed most clearly

by integrated displacements. Estimates of the total rupture duration are 7 to 1 0 sec, but the most prominent components are shear waves of 5 to 7 sec travelling northeast with ground

motions parallel to the fault, and shear waves

of 3.5 to 4 sec travelling to the northwest and

southeast with ground motions transverse to the fault.

5. REFERENCES

1. Ward, P. L. , and R. A. Page; The Loma Prieta

Earthquake of October 17, 1989; U.S. Geological Survey Pamphlet ( 1 989 )

.

2. Plafker, George and J. P. Galloway (editors); Lessons Learned from the Loma Prieta,

California, Earthquake of October 17, 1989; U.S. Geological Survey Circular 1045 (1989).

3. Shakal, A. and others; CSMIP Strong Motion

Records from the Santa Cruz Mountains (Loma

Prieta) California, Earthquake of 17 October

1989; Report No. OSMS 89-06 (1989).

4. Maley, R. and others; U.S. Geological Survey

Strong-Motion Records from the Northern

California (Loma Prieta) Earthquake of October 17, 1989. U.S.G.S. Open-file Report

No. 89-568 (1989).

5. Brady, A. G.; History of Accelerogram Data Processing Proceedings, Golden Anniversary

Workshop on Strong-Motion Seismomatry, USC, Los Angeles ( l 984 )

.

6. Trifunac, M. D. , F. E. Udwadia and A. G. Brady; Analysis of errors in digitized

strong-motion accelerograms; Bull. Seis. Soc

.

Am. , 63, 1 , pp 157-187, 1 973.

7. Fletcher, J. B. , A. G. Brady and T. C. Hanks,

Strong-Motion Accelerographs of the Oroville

Aftershocks: Data Processing and the Aftershock of 0350 August 6, 1975; BSSA, 70,

1 , p 243-267 (1980).

8. Lew, H. S. (editor); Performance of Structures During the Loma Prieta Earthquake

of October 17, 1989. NIST Special

Publication 778 (1990)

.

Amplifications and resonances at 1.5 sec are clearly responsible for some of the more impressive damage to the Bay Bridge.

TA8LE 1. INSTRUNENTS IN THE ARRAYS

Array Station Cooperative Long Network ?er<od(s;

1. Anderson Dam US6S

Crest, Left ADutment, Downstream

2. SHroy »1, I, 3, 4, 6, 7 CSHIP

3. Stanford 5LAC US6S Stan'orO Parking

Palo Alto VA Hospital

Henlo Park VA Hospital

«. San Francisco; Pacific Heights CSfllP San Francisco: Rincon Hill CSMIP

TerOa Buena Island CSHIP

Oakland: Outer Harbor Wharf CSHIP Emeryville USGS

RHS error

in a cc 1

.

(cm/sec^)

TABLE 2. DISPLACENENT ERRORS

RHS error In d 1 so 1 ac emen t (c«l

4 sec 6 sec 8 sFc 1 0 sec

2.5 0.261.0

0.1 2.4

0.24

4.2 0.426.6 0.66

<S

14 •! A 17

Figure 1. Location of 4 arrays. Other USGS stations and CSMIP stations in the arrays

are shown.

286
o

if) o \ r- o

n o

3 o

E<

CN

I O _

1 1 1—I I I I I "I I I—I

I I I I

I

1 1—I—I I I I I

I

1 1—I—I I r I Ir

I I I I I I I ' I

I

I I ' I I J I 1 I I I I I J I 1 'I '' 0.1 1.0

Frequency, Hz

10. 100

Figure 2. Fourier amplitude spectra of Anderson dam, left abutment, 2^43° component (upper)

and a reference trace processed identically (lower); no long-period filter, subtraction of fixed trace, and subtraction of mean.

12.0-1

Figure 3. Anderson Dam, 2^1° components, at crest (upper), downstream and left abutment

(lower); Y-direction offsets are for clarity, X-direction offsets align the long periods.

287
12.70

SECONDS

longitudinal direction orthe Bay B^^dge.

<^^^-<=txon. approximately in the

288
The Loma Prieta Earthquake: Site Response and Liquefaction

by

Ellis L. Krinitzsky

A. G. Franklin^

ABSTRACT

During the magnitude 7.1 Loma Prieta earthquake of

October 17, 1989, damage was widespread in the

mountains and towns near the epicenter, but otherwise was concentrated in packets around the margin

of San Francisco Bay. These pKDckets occurred where

vulnerable structures were founded on deposits of

bay mud, which amplified the ground accelerations

in the east-west direction by an average factor of

2.8, and in the north-south direction by 4.2.

Occurrences of liquefaction outside the epicentral

area typically were in artificial fill materials

placed on bay mud. The combination of thick bay

mud deposits and liquefiable sandy fill was particularly devastating in the Marina District of San

Francisco

.

KEYWORDS: Earthquake; liquefaction; site response

1. INTRODUCTION

The Loma Prieta earthquake was a magnitude 7.1

event that occurred at 5:04 p.m., California

Daylight Saving Time, on October 17, 1989. The

earthquake is believed to have been generated by

movement on a segment of the San Andreas fault with

"sympathetic" activation of branch faults. However, there is no evidence of relative movement on

the surface trace of the San Andreas fault. The

epicenter, principal active faults in the region,

and the area of microearthquakes that were recorded

by the U.S. Geological Survey within two days

following the event, are shown on Figure 1 (from

Plafker and Galloway, 1989). The microearthquakes

indicate the extent of fault rupture in the subsurface. The inclusive area that is represented is

approximately 150 km by 60 km and includes all of

the metropolitan area from San Jose to San Francisco and Oakland. Within this area there was appreciable but localized damage to structures from

shaking, soil liquefaction, and ground displacements .

2. EXTENT OF DAMAGE RELATED TO LIQUEFACTION

Earthquake effects were widespread in the mountains

and towns near the epicenter and in small isolated

pockets throughout the San Francisco Bay area. The

greatest property damage occurred in the towns of

Santa Cruz and Watsonville and in the Marina

district of San Francisco. There were numerous

fatalities, pronounced highway damage, and dislocation of traffic principally in Berkeley and Oakland

on InterBtate-80 and Interstate-SSO, closure of the

San Francisco Oakland Bay Bridge, destruction of

the bridge at Struve Slough on Highway 1 at Watsonville, and extensive damage to Highway 17

north of Santa Cruz. Additionally, there was runway damage at the Alameda Naval Air Station and

the Oakland International Airport, damage in the

terminal building of the San Francisco International Airport, plus damage to property and roads from

earth movements and landslides throughout the Santa

Cruz mountains. Minor landslides occurred along

the Pacific Coast, posing potential hazards where

Highway 1 and urban developments lie in close

proximity to the coast, as at Half Moon Bay,

Pacifica, and Pescadero.

Liquefaction effects were observed around the

margins of San Francisco Bay at several locations

on bay muds, especially at sites where sandy fill

had been placed over the bay mud. Such occurrences

include the San Francisco Marina District, the

Oakland International Airport, the Alameda Naval

Air Station, and the Port of Oakland. Outside the

Bay area, liquefaction occurred in young alluvial

deposits near the epicenter, notably on the Pajaro

and San Lorenzo Rivers.

2.1. San Francisco - Oakland Area

In the San Francisco - Oakland area, there were

disruptive and costly damages to bridges, freeways,

wharves, buildings and the major airp>ort8. The

center of this area is about 90 km from the epicenter of the earthquake, a distance at which the

ground motions due to the earthquake should have

been greatly attenuated. Reference to Figure 1 shows this area as Modified Mercalli (MM) intensity

zones VI and VII. Within these zones are several

intensities VIII and IX that are in the form of

localized, anomalous intensities. The MM intensities in this range may be generalized as:

VI Essentially no structural damage

VII Negligible structural damage

VIII Threshold of damage to good construction

IX Significant damage to good construction

The anomalous intensities in Figure 1 are clearly

the result of locally unfavorable conditions euch

as bay muds and artificial fill; all of the anomalies are in such areas. Figures 2 and 3 show

thicknesses and depths of younger bay mud; figure

4 shows the depth to bedrock. In figure 4, the

Marina District is shown as a deep embayment. The

contours on figures 2 to 4 are from Goldman (1969).

The bay muds are compMDsed mostly of marine clays

that are intermixed with lesser quantities of silts

and fine sands. Additionally, there is organic

U.S. Army Engineer Waterways Experiment Station,

Vicksburg, Mississippi 39180

289
matter including some peat. Upper layers of the

mud are unconsolidated and may be semi-fluid. The

Marina District is one of the areas in which rubble

from the 1906 San Francisco earthquake was dumped

and later covered with sand.

2.1.1. The Interstate - 880 Collapse

A section of 1-880 in Oakland, known both as the

Nimitz Freeway and as the Cypress Avenue Expressway, suffered a collapse in almost 2 km of the

upper deck during rush hour traffic. The earthquake motions caused the reinforced concrete

columns to crack leaving the hinge joints supporting the decks unstable. As the columns buckled,

the decks slid and dropped.

area is in the center of the picture. On an opposite corner is a building that was twisted off

its foundations. In the rest of the area the

damage is generally less spectacular though on the

ground one can see much damage in the first stories

and in the pavements and streets.

Figure 7 shows the boundaries of the area in which ' major damage occurred in the Marina District. As ' seen in figure 4, the area is underlain by deep H deposits of bay mud. On top of the bay mud is ' artificial fill over which sand has been placed, "f^ Additionally there are areas of naturally occurring '] sands from beaches and dunes. Since the area is

nearly at sea level and essentially flat, much of

the near-surface sand was water saturated.

Damage to the freeway is almost the only damage in

that area. The buildings are mostly intact, as are the other elevated sections of highway in the area. A few collapsed chimneys and brick facings were observed. The 1-880 collapse occurred entirely in

a segment that was founded on bay muds. Horizontal

motions in this area for rock, alluvium and bay mud

during a magnitude 4.1 aftershock (Borcherdt and

Hough, 1989) show that the bay mud had the effect

of amplifying the motions by a factor of about 5 compared to rock and about 2 compared to alluvium.

Amplification of motions in the bay mud also has a spectral effect observed in longer predominant

periods of motion.

Earthquake motions recorded in the free field on bay muds nearby at Emeryville during the Loma

Prieta earthquake had a peak horizontal acceleration of 0.25 g.

2.1.2. The Bay Bridge

A single span of the San Francisco - Oakland Bay

Bridge collapsed during the earthquake. The bridge

was otherwise structurally intact. Lateral move- ment on the order of several feet appears to have

occurred in the piers. The movement sheared bolts

in the upper structure holding the support beams,

thus permitting the upper deck to crash down.

Again the culprit appears to be in the greater

excitation that was generated in this portion of

the foundation soils. The bay muds have a thickness of approximately 40 ft according to the

interpreta.tion in figure 2.

In aerial photographs made the day after the

earthquake (figure 5), sand extrusions characteristic of liquefaction can be seen on the approaches

to the toll plaza on the Oakland side of the Bay

Bridge. These deposits of sand were smaller in

extent, but otherwise similar to, those seen on the

runways of the Alameda Naval Air Station and the

Oakland International Airport.

2.1.2. The Marina District

The Marina District in San Francisco is an area that suffered a concentration of costly damage that

included a fire resulting from a burst gas main.

Figure 6 shows the Marina area on 18 October 1989,

the day following the earthquake. The burnt-out

The combination of foundation conditions was ideal

to assure serious damage in a strong earthquake:

a deep, soft soil column to amplify the bedrock

motions; dumped fill and liquefaction-susceptible

sand overlying it; and a high water table to assure saturation of the sand.

Figure 8, from Plafker and Galloway (1989), shows

the effect of soil conditions in amplifying ground

motions during a magnitude 4.6 aftershock of the

Loma Prieta earthquake. The accelerograph locations are shown in figure 7. The vertical velocities were measured on sandstone, dune sand, and

artificial fill. The water-saturated dune sand

shows peak motions more than 5 times as great as in

the adjacent sandstone outcrop.

Some effects of soil liquefaction are seen in

figures 9 to 11. Figure 9 shows sand extrusion

with blow holes in the Marina Green. Figure 10

shows sand extrusions which ran out from under a garage door and over the sidewalk. Numerous cases

of sand extrusions into basements were reported.

Lateral translation of the pavements riding on liquefied sand caused the buckling and breakup of

the sidewalks and streets in figure 11.

2.1.3. The Port of Oakland

The port facilities in Oakland suffered considerable damage from shaking and from soil liquefaction. Damage to the wharves occurred from shaking

of the pile foundations whereas liquefaction was

the chief cause of damage in the adjacent soil

areas. Liquefaction was observed at distances of

up to 35 m inland. Pavements were dropped in those

areas relative to the docks by up to 45 cm.

2.1.4. Alameda Naval Air Station and Oakland

International Airport

Runways at the Alameda Naval Air Station and at the

Oakland airport suffered spectacular effects from

liquefaction. Figure 12 shows sand extrusions on the runways at the Alameda Naval Air Station, one

of which, in the center of the photo, is 120 m in

length and 30 m in maximum width. There is a linearity in the alignment of these liquefaction

occurrences. The soils were placed as fill materials. In the areas marked with crosses considerable

cracking took place in the pavements. Damages in

290
the pavements at the Oakland airport were similar.

2.2. Santa Cruz

The experiences in the Marina District of San

Francisco, where damage resulted from a combination

of liquefaction in the foundation soils and weaknesses of unsuitably built structures, were repeated in other areas. The business district of Santa

Cruz was severely damaged as a result of unfavorable construction, in this case unreinforced

masonry, ground motion amplification in alluvial

soils, and possibly liquefaction in the foundations.

The depth of soft soils varies from 7 to 165 m and

contains very young alluvial deposits. The area of

severest property damage is adjacent to the present

course of the San Lorenzo River. Some deposits of

soil which might have been extruded by liquefaction

were observed, but they could have resulted from

broken water lines. No gross distortion of pavements, such as that seen in the Marina District,

was observed here.

Santa Cruz was close to the epicenter of the

earthquake and experienced peak motions that were measured nearby at 0.60 g vertical and 0.54 g horizontal.

Figure 13 shows fissures and undulating displacements in the levee along the San Lorenzo River near

the business district of Santa Cruz. Figure 14 shows the pattern and extent of these fissures.

The levees were severely damaged throughout their

lower reaches. Ground displacements were mostly in

the form of elongate fissures that paralleled the

length of the levees. Eyewitness accounts of the

earthquake describe large ground waves that moved

through this area of very thick, unconsolidated

alluvial deposits in which the water table is near

the surface. After the earthquake, sand extrusions

adjacent to the levee were apparent over much of

its length, indicating liquefaction of the foundation soils under the levees.

2.3. Watsonville

Downtown Watsonville had a concentration of damage

in the business district similar to that which

occurred in Santa Cruz. The causes were the Seune; unreinforced masonry construction on very thick,

unconsolidated soils with a high water table

resulting from the local situation in a valley

along the Pajaro River. There was also widespread

damage in residential areas, mostly due to collapse

of chimneys and inadequate lateral bracing at the

foundation connections.

Levees on the Pajaro River in and near Watsonville

were deunaged extensively, much as in Santa Cruz. Figure 15 shows elongate fissures in the levees

downatreaxn from Watsonville. Some of these fis- sures were measured at 2.0 to 2.3 m maximum depth,

which makes them slightly deeper that the levees

are high. Sand extrusions in a plowed field

adjacent to the levee are shown in figure 16. There is a linear rupture with extrusion of sand

and multiple vent holes along the alignment. Other occurrences of sand extrusion were common in the

vicinity of the levees.

3. EARTHQUAKE GROUND MOTIONS

Figures 17 and 18 compare peak horizontal accelerations at hard and soft sites respectively as measured during the Loma Prieta earthquake with the

Krinitzsky-Chang-Nuttli (1988) curves for aM * 7.0

earthquake. The data were obtained from reports of

the U.S. Geological Survey (see Maley and others,

1989) and the California Division of Mines and

Geology (see Shakal and others, 1989). The curves

are for mean and mean plus one standard deviation.

The distances shown are focal distances.

A hard site, as distinguished from soft, has a shear wave velocity greater than 400 m/sec and

surficial soft cover is less than or equal to 15 m. Values at hard sites from the Loma Prieta earthquake (figure 17) are in general agreement with the

Krinitzsky-Chang-Nuttli curves.

Values at soft sites have a much greater spread in

values, almost a full order of magnitude at a focal

distance of 80 )tm. The hard sites have half of

this spread. Figure 18 indicates which values are

for bay muds. Eight of the nine bay mud locations

provided records that have peaks well above the mean plus one standard deviation for magnitude 7. Anomalously high values such as these were observed

for motions in the soft lake deposits in Mexico

City during several earthquakes. Amplification

factors there were as much as 5 times for soft versus hard ground (Krinitzsky, 1987).

Figure 19 compares Loma Prieta accelerograph

records for a granite and a bay mud site at roughly

similar distances, 52 versus 67 km, from the

earthquake source. The amplification factor for

the bay mud in this instance is about four times.

There is also a change in the spectral composition

of the bay mud record whereby there is a shift to

a longer predominant period, in this case about two

seconds. This effect is further illustrated by the

data in Table 1, in which ground motions at hard

sites are compared with those at several soft sites

with comparable focal distances.

Compared with hard sites, peak horizontal accelerations at bay mud sites were amplified an average of

2.8 times in the east-west direction and 4.2 northsouth. An extreme amplification of 9.7 times

(north-south) occurred in Oakland harbor.

4. CONCLUSIONS

Outside of the epicentral region, severe damage and

liquefaction during the Loma Prieta earthquake

occurred in localized pockets in which the Modified

Mercalli intensity was as much as two levels higher

than in the surroundings. These anomalous pockets

of damage were characterized by the existence of

vulnerable structures founded on deposits of bay

mud, in which the ground accelerations were amplified by an average of 2.8 times in the east-west

291
1

direction and 4.2 times in the north-south direction. Occurrences of liquefaction were also

characterized by the presence of sandy artificial

fill overlying the bay mud. The Marina District of

San Francisco, the San Francisco - Oakland Bay

Bridge, the Cypress Expressway on 1-880, the Port

of Oakland, the Alameda Naval Air Station, and the

Oakland International Airport were affected.

Existing distance-magnitude-motion curves do not account for the high values of ground motions at these sites, although similar occurrences have been

observed on very soft sites during past earthquakes, notably in Mexico City.

5. ACKNOWLEDGEMENT

The work reported in this paper was carried out under funding provided by the U.S. Army Corps of

Engineers, and is published by permission of the

Office of the Chief of Engineers.

6. REFERENCES

Borcherdt, Roger D., and Hough, Susan E. 1989. "Mud

May Have Contributed to the Collapse of Niraitz

Freeway during the Loma Prieta Earthquake," Earth

in Space . Vol. 2, No. 4, pp 6-8.

Goldman, Harold B., ed. 1959. Geologic and Engineering Aspects of San Francisco Bay Fill , Special

Report 97, California Division of Mines and Geology, San Francisco, California.

Table 1. Amplification of ground motions

due to local site conditions

HARD SITE

Focal

Dist

km

Accel

g EW NS BAY MUD SITE

Focal

Dist

km

Accel

g EW NS

Amplificatior

EW NS

Hayward CSUH 71 .08 .08 San Francisco Airport 79 .33 .24 4.13 3.00

Foster City:

Redwood Shores 63 .29 .26 3.63 3.25

Redwood City 63 .23 .28 2.88 3.50

Foster City 66 .11 .12 1.38 1.50

Yerba Buena Is. 95 .07 03 Treasure Is. 98 .16 .10 2.29 3.33

Oakland Wharf 95 .27 .29 3.86 9.67

San Francisco 18-Story 95 .14 .17 2.00 5.67

Berkeley, Strawberry 98 .08 .04 Emeryvi lie 97 .26 22 3.25 5.50

Canyon Larkspur 115 .14 10 1.75 2.50

AVERAGE AMPLIFICATION: 2.8 4.2

Krinitzsky, E.L. 1987. "Empirical Relationships for Earthquake Ground Motions in Mexico City," in The Mexico City Earthcruakes - 1985, Factors Involved

and Lessons Learned . American Society of Civil

Engineers, New York, 474 pp.

Krinitzsky, E.L. , Chang, Frank L. , and Nuttli, Otto

W. 1988. "Magnitude-Related Earthquake Ground

Motions," Bulletin of the Association of Engineering Geologists, Vol XXV, No. 4, pp. 399-423.

Maley, R. , Acosta, A., Ellis, F., Etheredge, E.,

Foote, L. , Johnson, D. , Porcella, R. , Salsman., M.,

and Switzer, J. 1989. U.S. Geological Survey

Strong-Motion Records from the Northern California

(Loma Prieta) Earthquake of October 17, 1989 , U.S. Geological Survey Open File Report 89-568, Menlo

Park, California, 85 pp.

Plafker, George, and Galloway, J. P., eds, 1989.

Lessons Learned from the Loma Prieta, California.

Earthquake of October 17. 1989 , U.S. Geological

Survey Circular 1045, Washington, D.C. 48 pp.

Shakal, A., Huang, M. , Reichle, M. , Ventura, C,

Cao, T., Sherburne, R. , Savage, M., Darragh, R.

,

and Petersen, C. 1989. CSMIP Strong-Motion Records

from the Santa Cruz Mountains (Loma Prieta),

California. Earthquake of 17 October 1989 . California Division of Mines and Geology, Report OSMS

89-06, Sacramento, California, 156 pp.

292
Figure 1. Location map for the Loma Prieta earthquake, showing Modified Mercalli intensities from

Plafker and Galloway (1989) and anomalous intensities noted in the San-Francisco - Oakland Bay area.

293
\

- BAY

BRIDGE

FAILURE

EMERYVILLE ^V^l FREEWAY Y

' FAILURE]

a4

OAKLAND

THICKNESS OF

YOUNGER BA Y MUD

CONTOURS 20 FT

SAN FRANCISCO / /( /^//Jt^V /

CYPRESS

EXPRESSWAY

DECK FAILURE

2 Ml KM

Figure 2. Thicknese of the younger bay mud with locations of Cypress Expressway deck failure, the Emeryville Freeway failure and the Bay Bridge failure. "—• -

> » Contours are from Goldman (1969

BAY BRIDGE

FAILURE

EMERYVILLE

FREEWAY

FAILURE

' OAKLAND

BOTTOM OF

YOUNGER BA Y MUD

CONTOURS 20 FT

SAN FRANCISCO i iy^lll/iL,,''^

74^ CYPRESS

EXPRESSWAY

DECK FAILURE

2 Ml KM

//,:<\

J^'" ~ ^ - -f^-"

«L*MCOA —^ . ^ ~>! V

Figure 3. Bottom of younger bay mud with locations

of Cypress Expressway deck failure, the Emeryville

Freeway failure and the Bay Bridge failure.

Contours from Goldman (1969).

294
ii

Figure 4. Top of bedrock with locations of the

Marina District in San Francisco, the Cypress

Expressway deck failure, the Emeryville Freeway

failure and the Bay Bridge failure. Contours from

Goldman (1969).

295
Figure 5. Aerial view of the toll plaza on the

Oakland side of the Bay Bridge, showing sand

extrusions on pavements. Taken on the day following

the earthquake.

296
I

I

297
MARINA DISTRICT

EXPLANATION

1 ARTIFICIAL FILL 2 BEACH DEPOSIT

3 DUNE SAND

4 ALLUVIUM

Figure 7. Limits of major damage in the Marina

District, and types of underlying soil deposits.

MARINA AREA: M = 4.6 AFTERSHOCK

3

10

^.VA',^<juvx-.v ... . MAS

SANDSTONE

20

PUC

DUNE SAND

LMS

FILL

30

Figure 8. Vertical velocities on sandstone, dune

sand, and fill in the Marina District during a M = 4.6 aftershock (Plafker and Galloway, 1989).

298
I

Figure 9. Sand boil from sand liquefaction in the

Marina Green, San Francisco.

Figure 10. Sand extrusion flowing out of garage

and over sidewalk, in the Marina District.

299
Figure 11. Buckled and broken sidewalks and street

pavement resulting from soil liquefaction, in the

Marina District.

300
301
Figure 13. Cracking and slumping of levee along

the San Lorenzo River in downtown Santa Cruz,

resulting from liquefaction of foundation soils.

LIQUEFACTION AT LEVEES, MAPPED 20 OCT, 1989

SANTA CRUZ. CA

Figure 14. Patterns of fissures in the levees of

the San Lorenzo River in downtown Santa Cruz.

302
Figure 15. Liquefaction-induced fissures in levee

of the Pajaro River, near Watsonville. The fis- sures were deeper than the height of the levee.

Figure 16. Liquefaction in a field adjacent to the

Pajaro River levee near Watsonville. Sand extruded

from a linear ground break, with multiple vent

holes

.

303
gOaS/I^O 'NOIlVd3n300V dOH

N ID H N

U 4J O --I £ C

Id ^

01

0) *

§•2

. 0)

0)u

01 u

U IM o

IM eg>u

3o

3Z

I u 18 a* 3

<a x:

lb 4J u

LUO

z

<

CO

o

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I I C >i o >: N O

O£ C

is s

i* ° o H o «

. 0) - s

0) o u «

3

-4 «

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304
GRANITE

MONTEREY - CITY HALL

(CSMIP STATION 47377) FOCAL DISTANCE 52KM

90*-

Up -

360'-

CODE 377

3 4 5 10

Uax. Acce 1

.

0.07 g

0 03 g

0.07 g

15 20 Sec.

BAY MUD

FOSTER CITY - REDWOOD SHORES

(CSMIP STATION 58375) FOCAL DISTANCE 67KM

90*'

Up

^

360*

CODE 376

Uax. Acce 1

0. 29 g

Oil g

0. 26 g

3 4 5 10 15 20 Sec.

LOMA PRIETA EQK, 1989

Figure 19. Comparieon of Loma Prieta accelerograph

records for a granite and a bay mud site.

305
Dam Performance During the Loma Prieta Earthquake

of October 17, 1989

Robert B.

ABSTRACT

Several dams were damaged and numerous others

had records of strong ground motions as a result of the October 17, 1989, magnitude 7.1 Loma Prieta earthquake. The tabulated

results of a survey on damage and recorded

ground motion are presented. Maximum dam

sections and locations of strong motion

instruments are shown where possible.

REWORDS: Dam; earthquake; earthquake damage;

Loma Prieta earthquake; strong motion

1. INTRODUCTION

Shortly after 5:00 p.m. on October 17, 1989,

the Loma Prieta earthquake (Ms - 7.1) initiated along the San Andreas fault. Shaking

from the earthquake was felt over a wide

region in west-central California. About a dozen embankment dams near to the epicenter

suffered strong shaking; numerous others

received only minor shocks. The strong

motion phase of shaking lasted only about 12 seconds . There are about 100 dams of various

sizes, most of them embankment dams, in a 100-kilometer radius from the epicenter.

Considering the numerous dams in the vicinity, damaged dams were few in number. Numerous strong motion records were recovered from

dams in the area; however, some of the severest shaking was not recorded due to lack of

instruments. This paper describes the per- formance of both damaged and undamaged dams.

Selected maximum dam sections and locations

of strong motion instruments are illustrated

where possible. Physical data and performance are tabulated in Tables 1, 2, and 3.

Not all as-built changes from original

designs may be shown in the dam sections. In addition, not all instrument records, mapping

of cracks and possibly related geologic

features, foundation information, and dam

design were available at the time of this

writing (January-March, 1990).

Mac Donald^

2. ANTICIPATED TYPICAL PERFORMANCE OF DAMS

Embankment dams are more numerous than con- crete dams, hence more records of embankment

performance during earthquakes are available.

After strong shaking, many embankments

exhibit settlement and longitudinal cracks in the crest area. Transverse cracks near to abutments are also common. Changes in seep- age and piezometer levels or pressures are not unusual, and over a period of time these

earthquake - induced changes often return to "normal .

In general, embankments built of compacted

clayey materials, when on a sound foundation,

appear to be fairly resistant to earthquake

damage. Conversely, embankments of poorly

consolidated sands or silts, tailings dams,

and dams qj\ liquefiable foundations are highly susceptible to damage.

The record of performance during earthquakes

for concrete dams is limited compared to embankments. However, all principal types of

concrete dams, including arch, multiple arch,

gravity, and buttress, have experienced high

peak accelerations, and all types have had at least one example of serious cracking (the

arch without earthquake loading). Improve- ments in design for seismic loading probably

would have prevented much of the known con- crete dam damage. Foundation conditions

appear to have been satisfactory in those

concrete dams experiencing damage.

3 . RECORDED PERFORMANCES OF DAMS AFFECTED BY THE LOMA PRIETA EARTHQUAKE

The locations of selected dams with some record of performance or close proximity to the earthquake are shown in Figure 1. The names, types, heights, and epicentral disU.S. Bmreau of Reclanation

Code D-3610, Building 67 DFC, Dorver, CO 80225

306
cances of these selected dans, are tabulated

in Table 1, while damage is tabulated in

Table 2.

3 . 1 Villiams Dam

w'illiams Dam was 10 1cm from the earthquake

epicenter, and this is the shortest epicentral distance for any dam of moderate to large size. The dam is a 21 meter-high

concrete gravity dam with a crest length of

27 m and a crest width of 3 m. The upstream

face is vertical and the downstream face

curves from the crest to the 19 meter-wide

base. The dam is on jointed, but sound,

conglomerate and it is only 0.6 km from the

surface trace of the San Andreas fault. The

reservoir is basically silted up and is not

fully performing its original function of

water storage.

The dam must have experienced strong shaking.

It did not have strong motion instruments.

There was no damage to the dam; however,

shallow rock falls occurred on the abutments.

3 . 2 Austrian Dam

Austrian Dan, Fig. 2, is a 56 meter-high

embankment located 12 km from the earthquake

epicenter. Austrian Dam was originally

designed as a zoned embankment, but was constructed as a homogeneous dam. It had

more damage than any other dam affected by

the Loma Prietra earthquake. During the

earthquake, the embankment settled up to

85 cm over the right two thirds of the dam,

while the right end of the dam moved 37 cm

downstream. A maximum upstream movement of

15 cm was recorded at the left quarter point

of the crest. Longitudinal cracks up to 4-1/2 meters deep formed on the upper 25 percent of the upstream dam face, while shallower longitudinal cracks covered much of the

downstream face. Transverse cracks occurred

in the embankment near both abutments, and

were the only cracks in the dam crest. Right-side transverse cracking near the

spillway was 4-1/2 meters deep. It is felt

that this transverse cracking is more severe

than would be expected for the amount of

settlement.

On the right abutment, a reinforced concrete

spillway, with its left wall in contact with

the embankment, suffered severe to minor

cracking throughout its entire length (headworks, chute, and terminal structure). Due

to tension cracks transverse to the spillway

axis as well as downstream movement, the

entire spillway structure seems to have

elongated by about 30 cm. Excavations along

the left spillway wall revealed that at least

part of the spillway chute and spillway wall

separated from the foundation and the embank- ment probably because of side-to-side rocking. In addition, some voids were created

under the upper spillway floor (as determined

by drilling) . The 8 meter-high spillway wall

may have accentuated the rocking.

The embankment and spillway were thoroughly

investigated by surface mapping, trenching,

and drilling, and by checking piezometers and

surface measuring points. Coincidentally

,

measurement surveys had been run the day

before the earthquake. Dam settlement,

lateral movements, and piezometric changes

due to the earthquake are shown in Figs. 2 and 3

.

Significant rises in the water levels,

Fig. 2, in four piezometers in the dam were noted after the earthquake. Piezometer No. 6 rose to above the existing reservoir level.

The water levels were trending back to "normal" by late October; however, subsequent

construction activities on the dam damaged

some of the piezometers and further water level observations are incomplete. The cause

of the water rise in the piezometers is not known, but it is speculated that embankment

pore water was expelled into the piezometer

pipes because of high pore pressures dynamically induced during the earthquake shaking.

The piezometers consist of slotted pipe with

a sand filter sealed in 5-foot intervals with

a 3/4- inch open riser pipe to the surface.

Rainfall was not a factor in the water level

changes, because the observations were made

several days apart when there was no rain.

Perched water is known to exist in the dam,

probably a consequence of layering in the embankment and former high reservoir levels.

Water was observed to have gushed out of the embankment in several of the post-earthquake

investigation trenches, and is generally

attributed to the perched water trapped in the layers.

Since the start of the rainy season was imminent, repair to the embankment and spillway were expedited. In early November a temporary repair of the spillway was initiated. Epoxy grouting was used in spillway

cracks where feasible; however, where cracks were too wide, Portland cement grout was used. Kany of the cracks have had only a surface treatment, but drill core samples

indicate good penetration of the epoxy grout.

The spillway damage will be reevaluated in

1990 and there is a potential of more repair

or even replacement.

307
Cracked embankment material and cracked

foundation near the spillway was excavated

and replaced by compacted embankment. Material to a depth of nearly 8 meters was removed

on the right and left sides of the embankment

near the abutments. Minor curtain grouting

was done in the left abutment.

A 1972 analysis estimated that 1-1/2 to 3 meters of settlement would result during

strong motion occurring during a magnitude M

- 8.3 earthquake. The maximum 85 cm of

settlement for a magnitude M - 7.1 earthquake

reasonably approximated the 1972 estimate;

however, the transverse cracking was more

severe than expected.

3 . 3 Almaden Dam

Almaden Dam, Fig. 4, is a 34 meter-high

embankment built in 1936 and located 15 kilometers from the epicenter. It has an upstream impervious zone covered with reinforced concrete paving and a downstream

pervious zone. Earthquake damage was negligible. Minor longitudinal cracking was found

in the upstream embankment face near the edge

of the concrete face paving. No repairs were necessary

.

3.4 Guadalupe Dam

The cracked berm material was subsequently

removed, replaced, and recompacted.

3 . 5 Newell Dam

Newell Dam, Fig. 7, is a 55 meter-high

embankment built in 1960 and located 19 km

from the epicenter. Newell Dam has a central

impervious core with an upper pervious and a lower random zone upstrefun and a random zone downstream. The spillway is on the left abutment. A sloping intake structure for the outlet works is along the left side

upstream dam face and connects to the former

diversion conduit in the dam base.

During the earthquake, longitudinal cracks

developed high in the upstream slope

(Fig. 7). Minor spalling occurred inside the sloping outlet works intake structure; in addition, there was local separation of the intake structure and embankment. Embankment

and foundation seepage measured at the downstream toe was found to have increased.

Since the earthquake, excavation and recom- paction of the cracked area has been com- pleted. The spalling that occurred in the intake structure will not be repaired at this

time. The foundation seepage has nearly

returned to normal.

Guadalupe Dam, Fig. 5, is a 43 meter-high

embankment built in 1935 and located 18 kilometers from the epicenter. It has an upstream impervious zone faced with reinforced concrete slabs and a downstream pervious zone. A modification was made to the

dam in 1972 by removing concrete slabs from

the upstream toe and placing a lightlycompacted, clayey buttress berm against the

upstream slope. This berm, not shown in Fig.

5, was placed up to elevation 570 feet, which

just overlaps the concrete slabs facing the

upper part of the slope.

The earthquake caused primarily longitudinal

cracking up to 2 1/2 cm wide on the upper 4 meters of the berm. These longitudinal

cracks had a maximvim depth of about 1.3 meters. By the time repair work was initiated, some of these cracks were 10 cm wide. Transverse cracks occurred in the embankment

crest near the right and left abutments.

Discontinuous minor longitudinal cracks

occurred on the crest and were confined to 60

cm downstream from a crest floodwall. Minor

cracking, spalling, and offset occurred in

the concrete panels on the upstream face.

Settlement amounted to 20 cm and there was

5 cm of transverse movement upstream.

3 . 6 Elmer J . Chesbro Dam

Elmer J. Chesbro Dam, Fig. 6, is a 29 meter- high embankment built in 1955 and located

19 km from the epicenter. Elmer J. Chesbro Dam is a compacted "homogeneous" earthfill

embankment with more -pervious materials

selected for upstream and less - impervious

selected for dowmstream. It has a 1.2 meter- thick riprap shell on the upstream slope and

a loose rockfill berm at the downstream toe. The rockfill in the berm is separated from

the earth embankment by a 1.2 meter-thick

gravel drain. The outlet works under the dam

is basically a 1.4 m (56 inch) I.D. welded

steel pipe in a reinforced concrete jacket

and placed on rolled embankment material.

The detached spillway is on the right

abutment.

During the earthquake, a 97 meter-long discontinuous longitudinal crack, up to 10 cm wide and 2.4 meters deep, occurred on the upstream shoulder of the dam crest. Vertical

offset along the crack was 6 cm down on the upstream side. The cause of the crack is attributed to settlement of the cohesionless

upstream riprap shell. Transverse cracks up

to 1.4 m deep occurred at the right end of

the crest. Transversely oriented cracks also

occurred in the natural ground of the

308
spillway approach near Co a steep shoulder of

the right abutment. An outlet works inspection after the earthquake disclosed cracks in

welds in two joints of the steel pipe; however, these cracks appeared to be of pre- earthquake origin. Crest settlement was up

CO 11 cm.

The cracked embankment was excavated and,

together with added material, replaced and

recompacted

.

3 . 7 Lexington Dam

Lexington Dam, Fig. 8, is a 62 meter-high

embankment built in 1953 and located 21 km

from the epicenter. Lexington Dam has a Chick General impervious core. The upscream

zone, originally intended to be a pervious

zone, is actually a random zone containing

impervious macerials, as allowed by a specificacion change. Downscream is a zone of

random semi -pervious materials. A narrow

filter zone with its base on a French drain

lies between the core and downstream semipervious zone. The outlet works extends

under the embankment. Bedrock consists

mostly of Franciscan sandstone and shale.

During the earthquake, transverse cracks

occurred in the embankment and natural ground

at both abutments. On the left side along

the downscream groin a 58 meter- long, discontinous, transverse, embankment crack system,

with individual cracks up to 2 cm wide aver- aged about 2 m deep. This cracking appears

to have been controlled by the steep abutment-embankment contacc. On the righc side cransverse cracks were boch upscream and

downscream of the crest, but were confined to natural ground (colluvium and highly weathered rock) . A continuous primary crack on

the right side from downscream Co upscream of

the crest reached a maximum width of 2 cm, averaged abouc 2 m in depch, and was about

137 meters long. Further upstream from this

primary crack, other cracks were discontinuous for another 85 m. Minor longitudinal

cracks were found on both upstream and down- stream faces of the embankment. Maximum

crest settlement was 26 cm and downstream

movement was 8 cm. An outlet works inspection afcer the earchquake disclosed a 9 mecer-long bulge in Che 1.5 meCer (50 inch)

steel pipe; however, the bulge was determined

to be of pre -earchquake origin. A Cemporary

rise in foundacion hydrosCatic pressure was reporCed.

Excavacion, replacemenc, and recompaccion of

Che cracked areas have been compleced.

The Lona Prieca earchqxxake shaking is welldocumented at LexingCon Dam by an excensive

array of scrong mocion insCrumencs. The

layouc and m^imum acceleracions recorded by

these instruaents are shown in Fig. 8. The

maximum crest accelerations cransverse and

parallel to che dam axis were 0.45g and

0.40g, respeccively

.

3.8 Vasona Percolacion Dam

Vasona Percolacion Dam, Fig. 9, is a 10 mecer-high eabankmenc, builc in 1935 and

locaced 24 kn from che epicencer. Vasona Percolation Dam has an upstream impervious

zone with a cutoff trench that is deep relative to the dam height. The upstream slope

of this impervious zone has a concrete facing. Several pervious to semi -pervious zones

are dovnstreaa.

During the earthquake, some transverse cracking occurred in the embankment near che right

abutment. The cracking pattern was primarily

longitudinal; cracks up Co 6 mecers long

formed along 146 m of Che righC embankmenc

cresc. All of che cracking was found co be relacively shallow (maximum depth - 1.7 m)

and the dam remained in a normal operation

mode

.

The cracked embankmenc was excavaced and

recompacced.

3.9 Rinconada Reservoir Embankmenc

The offscreaa embankmenc for Rinconada Reser- voir is 12 m in heighc and was builc in 1969. This reservoir is 26 km from che epicencer.

The embankmenc was undamaged during che earchquake. However, Che ouclec works, which

is locaced on nacural ground aparc from che embankmenc, incurred damage near ics incake

.

The walls of a vaulc, having a nozzle intake

from the reservoir, were cracked. The cracks

in the asphalt-coated concrete walls of the vault allowed water to pass from the vault

into pervious backfill.

The reservoir was drained and repairs co the walls of the vault have been completed.

3 . 10 LaroY Anderson Dam

Leroy Anderson Dam, Fig. 11, is a 72 mecer- high emfcankaenc builc in 1950 and locaced

27 km from che epicencer. Leroy Anderson Dam

has an upstream impervious zone, and semipervious and pervious zones downscream.

309
During the earthquake longitudinal cracks

occurred in the asphalt crest roadway. The

cracks may be related to a transition zone between the impervious zone upstream and more

pervious zones downstream that underlies the

crest. The cracks were about 1/2 cm wide at their maximum and had a maximum depth of

about 61 cm. This damage is very slight. Pavement sealing by patching has been completed.

The dam is well- instrumented to record strong

motion (Fig 11) . The maximum transverse

acceleration recorded on the crest was 0.43g

and at the toe was 0.23g.

3. 11 Covote Dam

Coyote Dam is a 42 meter-high rolled earth

and rockfill embankment built in 1936 and

located 31 km from the epicenter. There was

no damage at the dam from the earthquake

.

Coyote Dfun has, however, strong motion

instruments associated with it. A southwest

"abutment" site recorded a mfiximum acceleration of 0.49g. A downstream site had a maximum acceleration of 0.19g.

3.12 Cherry Flat Dam

Cherry Flat Dam is an 18 meter-high embankment completed in 1936 and located 42 km from

the epicenter. There was no damage at the

dam from the earthquake. An orthogonal selfcontained strong motion instrument was in

operation on the left abutment. The instrument recorded aaximum accelerations of 0.09g,

0.07g, and 0.06g in directions parallel to

the dam axis, transverse to the dam axis, and

up, respectively.

3.13 San Justo Dam

San Justo Dam, Fig. 10, is a 46 meter-high

zoned embankment completed in 1987 and located 46 km froa the epicenter. There was no

damage to the dam nor any found in the immediate area of the reservoir; however, measurements from survey points located along the crest of the dam indicate that up to 4 cm of

settlement occurred during the earthquake. A

temporary rise in foundation hydrostatic

pressure was reported.

The dam has an array of strong motion instruments (Fig. 10). Maximum axial acceleration

occurred in the center of the crest and was

0.39g while axial acceleration at the down- stream toe was 0.16g. Maximum transverse

acceleration was also at the center of the

dam crest and was 0.50g while transverse

acceleration at the toe was 0.26g.

3. U Del Valle da«

Del Valle Dam is a 68 meter -high embankment

completed in 1968 and located 66 km from the

epicenter. There was no earthquake damage in

the area of the dam. Strong motion instru- ments are located on the dam crest and at the downstream toe. Maximum axial accelerations

were 0.08g on the crest and 0.06g at the

downstream toe. The maximum transverse accelerations were 0.08g at the crest and

0.04g at the toe; the axial accelerations

were 0.08g at the crest and 0.06g at the toe.

3.15 Lower Crystal Springs Dam

Lower Crystal Springs Dam (Fig. 12) is a 43 meter-high concrete gravity structure com- pleted in 1888 and located 69 km from the epicenter. There was no earthquake damage in

the area of the dam. Strong motion recording

Instruments are on the crest, left abutment,

downstream toe, and downstream as shown on

the plan and section in Fig. 12. The maximum

crest acceleration was a transverse acceleration of O.lOg near the left abutment; the accelerometer in the center of the crest malfunctioned. At the downstream toe of the dam, both the axial and transverse maximum

accelerations were 0.05g.

4. CONCmSIONS

Considering the proximity of numerous dams to the epicenter of the major Loma Prieta

earthquake, very few of these structures were adversely affected. Eight dams or their

appurtenant structures sustained damage,

which ranged froa serious to very minor. The

greatest damage occurred at Austrian Dam where large settlements and cracks occurred

and the spillway was badly damaged.

There were two factors which contributed to the limited damage. The first was the short

period of strong motion shaking, only about

12 seconds. The second was the timing of the earthquake. The earthquake occurred in the

third year of below- average rainfall as well

as in fall when reservoirs were drawn down to low levels to satisfy water demands of the warm summer season.

Certain effects consistent In the performance

of embankment dams during earthquakes were documented. Longitudinal cracking and

embankment settlement, two common responses

to strong earthquake shaking at earthen and

rockfill structures, occurred at a number of

dams close to the epicenter of the Loma Prieta earthquake. Many transverse cracks

also occurred, and these features are of some concern to daa safety. Temporary

310
fluctuations In plezometrlc levels were docuoented for Austrian Dam and reported for

two other daos . Two concrete gravity dams

that are discussed had no damage.

5. AgtCNQUlEpgmarr?

The author Is appreciative of field notes and

inforaation from Don Babbitt, Dapartnent of Water Resources, State of California, and

gratefully acknowledges the Information and

reports provided by Robert Tepel of the Santa

Clara Valley Water District.

6. REFERENCES

1. Haley, R. et al , 1989, U.S. Geological

Survey strong-motion records from the

northern California (Loma Prleta) earthquake of October 17, 1989: U.S. Geological

Survey Open- file Report 89-568.

2. Melgh, A. C, 1980, Geotechnical

Invesigations for Dams, Proc. of

International Symposium on Geotechnical

Problems and Practice of Dam Engineering,

Asian Institute of Technology, Bangkok,

163-181.

3. Plafker, George, and Galloway, John P.,

Editors, 1989, Lessons learned from the Loma Prieta, California, earthquake of

October 17, 1989: U.S. Geological Survey

Circular 1045.

4. Shakal, A. et al , 1989, CSMIP strongmotion records from the Santa Cruz Mountains (Loma Prieta) , California earthquake

of 17 October 1989: Report No. OSMS 89-

06, California Strong Motion Instrumen- tation Program.

5. R. L. Volpe & Associates , 1990, Investigation of SCVWD Dams Affected by the Loma

Prieta Earthquake of October 17, 1989, Los

Gatos, CA.

6. USCOLD Committee on Earthquakes, Performance of Dams During Earthquakes, in preparation.

311
Table 1. List of Selected Dams

^9ni Noma Type Dam Height

ft m

Epicentral

Distance

mi kn

I ear CompleteDamage Strong

Motion

Williams Concrete

uravi ty

69 21 6 10 1895 None No

Ejnbankment 185 56 7 12 oevere WO

Almaden Embankment Tin "^4 q 1 s 1936 No No

Guadalupe Embankment 41 1

1

i. J. 18 1935 Yes No

MAUA 1

1

ilCWC 1 i. Elmb ankment i. w ^ 12 19 1 jOU I es INO

Embankment 95 99 12 19 I es

T OVl TICT^rtTl Embankment 90S 69 13 21 i. 7 J ^ I c s ICS

\7a V c j^nlaa Percolation

cjnDdnKmenc 34 10 15 24 i. ~ J J 1 CO

Rinconada

P A c a 1 T" CVCd V? L W \J 1. L Embankment 40 12 16 26 1969 Yes No

Leroy Anderson Embankment 235 72 17 27 1950 Minor Yes

Coyote Embankment 140 43 19 31 1936 No Yes

Cherry Flat Embankment 60 18 26 42 1936 No Yes

San Justo Embankment 147 46 29 46 1987 No Yes

Del Valle Embankment 222 68 41 66 1968 No Yes

Lower Crystal

Springs

Concrete

Gravity

140 43 43 69 1888 No Yes

312
Table 2. Earthquake Effects on Embankaents

Dam Name Cracks Maximum Maximun Other

Settlement Transverse

Longitudinal Transverse Movement

Austrian

Almaden

u/s & d/s

faces

in embank. 85 cm near both

abutments

minor on u/s none

enbank. face

not known

37 cm d/s

15 cm u/s

not known

piezometer level rise,

spillway cracking and

separation from foundation

Guadalupe

Newell

Elmer J. Chesbro

on u/s berm, in embank. 20 cm

minor along near both

crest abutments

high on u/s

slope

along u/s

crest

shoulder

none

at contact

of crest

and right

abutment

not known

11 cm

2 cm d/s concrete facing slabs u/s

5 cm u/s have minor cracking,

spalling, and offset

not known interior spalling and local

separation from foundation

of sloping intake structure

0.3 cm d/s crack in natural ground of

1.5 cm u/s left abutment along shoulder

of steep abutment

Lexington minor on u/s

& d/s faces

Vasona right side

Percolation embankment

crest

in embank,

left side,

in natural

ground on

rt. side

minor in

embankment

near right

abutment

26 cm

5 cm

7.6 cm d/s

4.3 cm u/s

3 cm d/s

none u/s

cracking near boathouse, near spillway bridge, along

spillway, and in natural

ground near intake structure,

foundation hydrostatic

pressure rise reported

Rinconada

Reservoir

none none not known not known cracked walls in separate

outlet works vault

Leroy

Anderson

San Jus to

intermittent none

along crest

none none

4 en

4 cm

2 cm d/s

1.5 cm u/s

not

significant

temporary peizometer water

level rise reported

313
Table 3. Maxlnnim Accelerations (g) Experienced at Instrumented Dams

Oaa Name Direction Crest Toe Other

Lexington Axial

Transverse

Vertical

0.40

0.45

0.22

None

Left Abutment

0.41

0.45

0.15

Leroy Anderson Axial 0.32

Transverse 0.43

Vertical 0.16

Left Abutment (rock)

0.18 0.08

0.23 0.07

0.16 0.05

Coyote Axial

Transverse

Vertical

0.17

0.19

0.10

Left Abutment

0.15

0.49

0.08

Cherry Flat Axial

Transverse

Vertical

Left Abutment

0.09

0.07

0.06

Embankment near Left Abutment

San Justo Axial 0.39 0.16 0.30

Transverse 0.50 0.26 0.24

Vertical 0.32 - 0.24

Del Valle Axial

Transverse

Vertical

0.08

0.08

0.07

0.06

0.04

0.03

Lower Crystal Axial None 0.05

Springs Transverse 0.10 0.05

Vertical 0.03

Left Abutment

0.07

0.09

0.03

314
Fig. 1. Location of selected dams affected by the October 17, 1989, Loma Prieta

earthquake with the earthquake epicenter shown. The earthquake was caused by ruptvire along the San Andreas fault. The epicenter is to the southwest of the fault since, in this area, the fault dips about 70 degrees southwest and the initiating rupture was at a depth of 18 ka.

315
W S. El iilO.O

20'0AM CREST El 1125.0

FOUNDATION, STREAM BED Ji

MATERIAL P -• PieZOMETER LOCATION

READINGS FROM NEW PIEZOMETERS

X PZ#lW«923') O PZ#tW(96d) O PZ#4W(950') V PZ *SW(9991

b.

18- MOV.

Fig. 2. Auscrian Dam. a. shows typical section andb. shows piezometer readings

before and after earthquake.

316
II3O.0

M22.0 ' ' I I 1 I I I I OJQO 2.00 4.00 6.00 8J00

STATION ALONG CREST

0.

1.2 r

ui ^

g 1^ ^

^ i 0.00 2.00 4.00 6.00 8 00

STATION ALONG CREST

b.

Fig. 3. Austrian Dam. a. Shows cujnulatlve settlement along dam crest Just

before and after October 17, 1989, earthquake. b. Shows horizontal

transverse movements of crest survey points just before and after

earthquake

.

317
20.0

2ao

"ig. 5. Guadalupe Dam. Typical section.

5.0 -CREST EL. 5 35

ALL UNITS IN FEET MAXIMUM NORMAL W S. El 525

RIPRAP {4 )

GRAVEL BLANKET ORAIN

45'

ROLLED EARTH FILL ^"^(MORE IMPERVIOUS I ^-^'"G"'-'- CSOUNO

(LESS IMPERVIOUS) LOOSE

ROCK

FILL

GRAVEL ALLUVIUM

EXCAVATED TO BEDROCK

UPSTREAM OF AXIS

Fig. 6. Elmer J. Chesbro Dam. Typical section.

318
Fig. 7. Newell Dam. a. Typical section, b. Plan of dam showing longitudinal

cracks along upper upstream face.

319
NORMAL w S. $50

CREST £1. 667

FOUNDATION

SEMI-PERVIOUS

20+00 18 + 00 16 + 00 i4-t-00 L_ 12 + 00

Fig. 8. Lexington Dam. a. Typical section. b. Plan of dan showing location

of strong>motion instruments and table of maximum accelerations of

s trong - mo t ion ins truments

.

320
lyffffVtOUS FIN€

kUreftlAL FROU BORROW PITS, W£T AND ROLLED

CONCRETE

FACIMt

ORIGINAL GROUND

MAximum w s. Ei. 300

20.0 *^EL303^ corpse MATERIAL FROti BORPOm PITS

MATERIAL FROm CJTOTf TRENCH

TOP OF IMfERVIOUS MATERIAL

CUTOFF TRENCH PUDDLED CLAY 25 25 SCALE OF FEET

50

ALL UNITS

IN FEET

Fig. 9. Vasona Percolation Dam. Typical section.

3

It

SMS

ACCELERATIONS

Sefwor

SM-I AXIAL VERTICAL TRANSVERSE SM-2 AXIAL VERTICAL TRANSVERSE SM-3 AXIAL TRANSVERSE Sy-4 AXIAL VERTICAL TRANSVERSE SM-8 AXIAL VERTICAL TRANSVERSE

Moxig)

0.39

0.32

0.50

0.33 0.17 0.2T

o.ie

0.26 0.30 a24

0.24 0.30 0.20

0.35 ALL ELEVATIONS ARE

IN METERS

MAXIMUM WATER SURFACE El I 53 .4

ORHilNAL GROUND SURFACEs

El. 127.0

I laO, CREST El. 155.0 —^ j £ ACCESS ROAD

ELECTED CLAY, SAND, SILT

TOP OF PLIOCENE

SEDIMENTS mSM-3

GRAVEL FILTER

Fig. 10. San Justo Dan. a. Plan view showing location of strong-motion

instruments and table of maxirniaa accelerations during the October 17,

1989, main earthqxiake event, b. Typical section, also showing location

of strong-motion instruments.

321
ACCELERATION

DIRECTION MAX(<jl

PLAN

^LEFT ABUTUENT

I WCK SITE SUA1

1ACCELERATION

OtRECTION MAX(4)

AXIAL 0.08 UP O.OS TRANSVERSE a07

ACXllXRATIOH

f

OOWNSTftEAM SUA1

ACCELERATION

DIRECTION MAXtq)

MO. Ort»€CT10H MAXt4l

1 AXUU. OJ 1 2 TRANSVERSE 0.14

3 TRMavCRSe 0.1* 4 AXIAL aie

S UP 0.16 « TTUMSVCRSe 0.23 7 AXIAL 0J2

8 UP O.I« 9 TmNsvotse 0.43

10 AXIAL 0.32

II UP 0.23

12 TMAMsvoee 03a

AXIAL O.ZS UP 0.13 TRANSVERSE 0.26

Fig. 11. Leroy Anderson Dan. a. Typical section, b. Plan of dam showing

locations of strong -mot Ion instruments and tables showing maximum

accelerations during the October 17, 1989, earthquake.

322
12 0.05

15 0.09

M 0.03

15 0.07

t.

Fig. 12. Lower Crystal Springs Dam. a. Typical section showing locations of

strong-motion instruments and table of maxijmin accelerations during

the main event of the October 17, 1989, earthquake, b. Plan of dam

showing locations of strong-notion instruments.

323
Perfonnance of Wood-Franied Stmctures in

the Loma Prieta Earthquake

by

John Tissell, P.E.^

ABSTRACT

The performance of wood-framed buildings in the Loma Prieta earthquake was investigated by a team of engineers from the Technical Services

Division of the American Plywood Association.

The purpose of the investigation was to determine if deficiencies in the building code were

a factor in building damage. All of the damaged

buildings examined were either built prior to the addition of seismic requirements to the code

or were constructed with obvious nonconformance

to building code requirements. Most failures were caused by lack of connections to the foundation, or lack, of lateral stiffness in the first floor or in short "pony" or cripple walls

between the foundation and the first floor.

KEYWORDS: Building Code; Connections; Earthquake; Shear Walls; Wood-framed.

1 . INTRODUCTION

In the late afternoon of October 17, 1989 a strong earthquake was felt throughout the San Francisco Bay area. The earthquake, subsequently known as the Loma Prieta earthquake,

registered 7.1 on the Richter scale. Much

publicity was given to the collapse of a section

of the double-decked Nimitz freeway in Oakland

where the majority of fatalities occurred, and

to the Marina district in San Francisco where a number of outdated residential buildings were destroyed, either by the earthquake or by the ensuing fire resulting from broken gas lines.

Other areas reporting heavy damage were Santa

Cruz County, location of the earthquake's

epicenter, and the town of Los Gatos, near the Santa Cruz Mountains. Farther north on the San Francisco peninsula, several homes in the Los Altos Hills area received heavy damage.

On October 23, a team of three engineers

from the Technical Services Division of the American Plywood Association traveled to the San Francisco Bay area to spend several days

observing the effects of the earthquake on wood construction. Those participating were Daniel H. Brown, John R. Tissell and William A. Baker. The primary objective was to determine

if deficiencies may exist in building codes

published subsequent to the Sylmar quake in the San Fernando Valley of California in 1971. The

Sylmar earthquake resulted in several building

code changes designed to strengthen buildings

against severe lateral forces.

The team discovered in short order that damage

was quite widely scattered. Vast areas between

the highly publicized damage zones appeared to have little or no structural damage from the earthquake

.

2 . CONVENTIONAL WOOD-FRAMED BUILDINGS

The epicenter of the earthquake was located in the Santa Cruz Mountains near the Loma Prieta

school. The quake opened up several ground

fissures of considerable length. The fissures were in the proximity of Summit Road and neces- sitated pavement repairs in several places

where they crossed the road. In this area, primarily populated with wood-frame residences,

the extremes of no apparent structural damage

whatsoever to complete destruction were observed. Even in areas where damage was high, many wood-frame "success stories" were encountered. The house pictured in Figure 1 appeared

to sustain no structural damage other than the collapse of the masonry chimney. The house was

two stories on a sloping lot, with a woodframed daylight basement to the rear and was sided with plywood panels which acted as bracing. It was located approximately 100 feet from a ground fissure, and undoubtedly was shaken rather severely. In fact, it was located about 200 yards from the house shown in Figure 8 which was completely destroyed in the earthquake

.

Dennis J. McCreary, staff engineer with the International Conference of Building Officials,

promulgators of the Uniform Building Code,

visited Santa Cruz County shortly following the earthquake. He observed, "It appeared that where there were failures in wood-frame buildings, they were due to inadequate connections,

or lack of structural continuity in design."

2.1 Cripple Walls

"Cripple" or "pony" walls appeared to have been

the source of failure in a number of damaged

homes. Cripple walls are typically stud walls

which spring from the foundation to the first

floor framing. They are often used on stepped

foundations in buildings built on sloping sites

and are usually short, but can be as much as full story height. When cripple walls have a

1. American Plywood Association, Tacoma,

Washington 98411

324
stud height exceeding 14 Inches, Che Uniform

Building Code requires that they be considered

first-story walls for the purpose of determining

the bracing required. The code also requires

that cripple walls exceeding 14 inches shall be framed with studs not less in size than that of the studding above and, when. they exceed 4 feet

in height, cripple walls shall be framed of studs having the size required for an additional

story. These requirements clearly indicate that such walls are required to provide resistance to lateral forces, and are not merely skirting.

Several instances were observed where houses had been moved laterally off tneir founaations. In every case of this type which was observed, the cripple walls were laying flat on their sides as shown in Figure 2. Although these walls were typically sheathed or sided with plywood panels,

nailing was typically much sparser than required

by the code. Thus, the panels themselves could

not provide sufficient lateral resistance to prevent racking of the cripple walls. No instance was observed where a cripple wall collapsed which was nailed in accordance with code

requirements

.

In the Los Altos Hills area, one house was observed where a rather long section was built over a garage on a sloping site. Nearly all the side walls of the garage, which supported the first floor, can be considered a cripple wall.

Two-foot-wide plywood walls (veneered with stone masonry) were built on either side of the garage

door opening. During the earthquake, this section of the house appeared to rotate coun- terclockwise, resulting in detaching the room over the garage completely from the rest of the house ana laying over the narrow walls beside

the garage door opening about one foot (see

Figure 3). At the location where the room detached from the rest of the house, there

appeared to be no structural continuity, as shown in Figure 4. Here, again, nailing of the sheathing along the cripple walls was less than

that required by code. In addition, the garage

door header, which supported a portion of the first floor, was not adequately tied to the sup- porting walls, further, the narrow walls beside

the garage door opening were not anchored with

hold-downs appropriate for shear walls. Still,

the room above the garage, complete with

unbroken picture window, moved as a unit and was undamaged except for where it separated from the rest of the house.

2.2 Houses Without Cripple Walls

In an older section of Los Gatos, many houses were observed that had slipped laterally off their foundations. The age of these houses

would indicate that they were board sheathed and

the foundations appeared to be unreinforced

masonry. Due to the age of the houses, and that

they were obviously constructed before modern

building codes, no detailed inspections were made by the team.

In the same area, the house shown in Figure 5 was seen. Even chough it has a double garage

in che ground floor, ic appeared undamaged.

This house appeared to be builc under recenc building codes.

The cwo-scory house wich basemenc garage shown

in Figure 6 showed no indicacion of damage even

though it was very near two houses that were

badly damaged (one of which is shown in Figure

12).

Figure 7 is a house near Summit Road and north

of Highway 17. Ground movement in the area is evidenced by buckled blacktop in the road near

the house and several broken buried water

lines. Even chough the entire end of the house

is Caken up by a double garage door, chis house

did noc show any indication of damage. There

was an approximately one-inch crack in the walkway between the street and the house, which

indicated shifting of ground in Chis area.

2 . 3 Major Framing Conneccions

Two of Che badly damaged houses Chat were examined showed very liccle conneccion becween che major floor framing members and che foundacion.

While boch of these houses had stepped foundations, each had one side where the floor framing bore directly upon the sill plate which was well bolted to che concrete foundacion.

In che reccangular house where the 2x floor

joists were on the sill plate, there was no evidence of fasteners between the joists and

che place. The Uniform Building Code requires

Chree 8-penny nails per joisC for Chis connection. The perimeter band joist was nailed to each joist; however, the band joist was fastened to the sill plate with toe nails

approximately 4 feet on center.

The second house was octagonal in shape, and

the floor was framed with 4-inch-wide beams

(see Figure 8). A careful examination of Che sill place showed ac Che inosc only one toe nail

connecting the floor joists to the sill plate.

Again, the floor framing members were nailed to the perimeter band joist and, in this case, the perimeter band joist was anchored to the foundation using hold-down anchors with lag screws

into Che band joist. The earthquake movement

was such that the lag screws failed in withdrawal from the band joist.

2 .4 Construction in the Marina District

The portion of San Francisco known as Che Marina discricc is an area where the rubble

from the 1906 earchquake was dumped and Chen

subsequencly filled with sand dredged from the

bay. During the 1989 earchquake the fine sand

liquified and, in many cases, flowed up through

cracks in the road or sidewalk. In other

instances, che sidewalks were badly discorced

by pressure from Che sand.

325
The typical building in Lhis area is a four-story apartment consisting of three floors

above ground floor garages. Around the most badly damaged buildings, inspection was possible

only by observations from outside the police

barricades

.

The large number of garage door openings on the ground floor of these buildings resulted in only

d very small length of exterior wall for racking

resistance. These walls were typically con- cealed by masonry veneer; however, in many

instances, the masonry had fallen away exposing

the sheathing. The sheathing was typically

horizontal boards and in the one building where

the sheathing had fallen off with the masonry,

there was cut-in 2x4 diagonal bracing. The

few interior partitions that were visible in the garages were sheathed with gypsum wallboard.

While several of these apartment buildings did

collapse, many more showed evidence only of distortion at the first floor. Figures 9 and 10 show one of these buildings. House movers were already at work placing shoring and steel beams

to prevent further collapse and to enable jacking up of the buildings and restoring the ground

f loor

.

The small, approximately 2-1/2-story apartment

building shown in Figure 11 was entirely

sheathed with plywood. On this particular

building the exterior damage was superficial,

with a small 2-foot-hlgh brick veneer wall

falling away from the building. Investigation

showed no ties between the brick and the building. Upward ground movement was sufficient to completely jam the garage door, so the owner

found it necessary to saw off the bottom of the door to open it after the quake. The concrete

garage floor was totally destroyed and will have

to be replaced, but the building above showed no evidence of damageAt another location, a contractor was adding

plywood sheathing to the interior of a garage to add lateral resistance to a relatively undaooaged

apartment building. The contractor pointed out

that there were no anchor bolts from the sill

plate to the foundation, so his first step was

to drill in anchor bolts. He then added block

ing between the studs and was applying 3/8" plywood horizontally to the walls.

One large apartment complex in the Marina

district was destroyed by fire initiated when

natural gas lines were broken due to the earthquake. This can also be related to lateral load resistance of buildings. According to Franklin

Lew, Manager of Seismic Safety Program for the city and county of San Francisco, "Wood-frame

buildings designed to provide lateral resistance

are less likely to sway and drift in an earthquake. This limits breakage of electrical, gas

and water lines within the building, thus reducing the chance for catastrophic fires."

3 . CHIMNEYS

The chimney damage shown in Figure 1 was common

to many houses whether or not they were further

damaged. The chimneys typically were not reinforced and were not tied sufficiently to the structure. Chimney failures were seen in many

of the areas observed. Some residences con- structed in the last 15 or 20 years have utilized insulated steel chimney flues enclosed

by decorative plywood siding. A number of these were observed, and none had collapsed

except for one which was not well fastened to the house and also was sheathed with heavy

stone veneer.

4 . POLE-SUPPORTED BUILDINGS

There were many pole-supported houses located

on steep slopes in the town of Los Altos Hills

- about 25 miles from the epicenter.

Some of these suffered extensive damage, and

the one shown in Figure 12 appeared to be a total loss. The damage seemed to be due to two factors: 1) inadequate connections between the floor system and the pole foundation, and 2) unequal stiffness of the poles.

The first deficiency caused the floor systems

to shift relative to the poles. This movement

was sufficient in some cases to cause the floor

framing to lose support from the poles.

The second deficiency was due to the unequal

height of the poles. Even if the floor system

successfully tied the poles together so they

acted as a unit, their varying stiffness meant

that the connections between the floor and the shortest poles carried the greatest lateral

forces during the earthquake. Those connections would be the first to overload and fail,

then the next shortest pole/floor connection

and so on.

Poles under some houses had been braced with

diagonal cables, and these structures appeared

to be undamaged. Repairs underway at one home

included the addition of diagonal cables.

5. INDUSTRIAL BUILDINGS

During the 197L Sylmar, California earthquake

there were numerous failures due to masonry and concrete walls pulling away from roof systems.

The failures were initiated by weak connections

between the walls and the roof framing. The Uniform Building Code was subsequently Improved

to correct this situation by requiring a positive tie between the walls and roof framing.

A small industrial building in Campbell, about

20 miles from the epicenter, had minor damage

of this type. A 4x12 timber ledger was bolted

to the masonry wall.

326

^ ^
The pl3rwood sheathing of the panelized roof was

Che only tie to the wall capable of resisting

outward (tension) forces acting on the wall. The nails holding the plywood to the top of the ledger held, but the ledger split along the bolt

lines. The splits in the ledger were apparently

caused by cross-grain bending. As mentioned

above, this detail is no longer permitted by the Uniform Building Code. In spite of the damage,

the plywood roof diaphragm functioned satisfactorily, preventing damage to the masonry walls.

6. MASONRY BUILDINGS

Masonry buildings typically perform poorly in earthquakes, particularly those that are unreinforced. Without reinforcement, masonry buildings fail in a brittle fashion. In the town of Los Gatos, about 15 miles from the epicenter,

an entire block of one - and two-story masonry

stores had been severely damaged and will

undoubtedly be demolished.

Unlike wood construction, which is light in weight, masonry is heavy. Its greater weight

leads to greater lateral forces to resist than

wood construction.

7 . SUMMARY AND CONCLUSIONS

A team of engineers from tne American Plywood

Association visited the San Francisco Bay area shortly following the Loma Prieta earthquake of October 17, 1989. Damage from the earthquake

occurred in scattered locations within the region. Within these locations, the heaviest

damage appeared to be to buildings with unreinforced masonry walls. All observed wood-framed

buildings which were damaged were either built

before the 1973 Uniform Building Code introduced

updated earthquake regulations, or incorporated

critical construction features but with inadequate connections, or other features which were

not nailed or stapled in accordance with minimum

code requirements.

It is the conclusion of the investigators that framed construction built to meet current code

provisions performed very well. No code

deficiencies for this type of construction were observed. However, greater attention must be paid to design and Installation of connections

and fasteners, and to structural continuity,

especially in small buildings which may not be engineered

.

327
Figure 1. Two-story house with wood-framed

daylight basement located about 100 feet from

ground fissure in Santa Cruz Mountains. No apparent structural damage except collapsed masonry chimney.

Figure 3. House in Los Altos Hills rotated

counterclockwise in portion over garage due to inadequate nailing of sheathing and connections

at narrow walls beside garage door opening. Rooms above appeared to move as a unit and remain sound

.

Figure 2. Collapsed cripple wall on house in Santa Cruz Mountains. Plywood sheathing on all cripple walls was inadequately fastened, allowing house to shift off foundation which split

the sill plate along the bolt line.

Figure 4. Break in house between kitchen and

recreation room. The kitchen floor was wood

framed over a crawl space. At the right end of the house, the crawl space became the garage. All of the major floor and wall framing members

had unspliced joints at the break.

328
Figure 5. Undamaged split level house located

in area of Los Gatos wnere many older

one— story houses slipped laterally off their

foundations.

Figure 6. Well constructed split level house

with no apparent damage. Two nearby single

level homes were damaged, one beyond repair.

Figure 7. Undamaged house in area of sufficient ground movement to buckle blacktop road

and break water lines.

Figure 8. Lateral displacement of floor at completely destroyed house in Santa Cruz County. The lag screws connecting the band joist to the anchors failed in withdrawal.

Other than a single toe nail, no fasteners were

found tying the floor joists to the sill, which

was bolted to the concrete foundation.

329
Figure 9. Corner apartment building in the Marina district. The masonry veneer was anchored to the sheathing by nails embedded in the mortar.

Figure 11. Plywood sheathed and sided

apartment building near the building shown in Figure 9. Except for the badly damaged con- crete garage floor, this building had minor

damage

.

Figure 10. Corner apartment building in the Marina district. House movers have shored the garage level in preparation for restoring the

bui Id ing

.

Figure L2. A badly damaged house in the Los Altos Hills area, supported on poles. The few connections between the poles and the floor

framing failed due to poor quality welds. The

floor joists were steel under the center por- tion of the house and wood at the edges. there

appeared to be a lack of continuity between the

two types of framing.

330
Seismic Performance of Nonstructural Elements

During the Loma Prieta Earthquake

by

T.T. Soong

ABSTRACT

For nonstructural elements, some of the national and

statewide (California) provisions have only recently been

applied to newly constructed facilities and to upgrading of existing nonstructural systems. The Loma Pri- eta earthquake provided a good opportunity for assess- ing whether these newly instituted standards have pro- vided added margin of safety to these systems under

strong earthquake disturbances.

Seismic performance of a selected class of nonstructural elements during the Loma Prieta earthquake is assessed

in this paper. Emphasis is placed on their performance

evaluation with respect to the criteria under which they were designed, upgraded or constructed. KEYWORDS: Building codes; earthquake resistant structures; ground motion; nonstructural damage; pres- sure piping.

1. INTRODUCTION AND BACKGROUND

Nonstructural elements of a building are generally considered as those materials that are not a part of the

structural system. They include equipment, architectural elements and building contents. The importance

of nonstructural element issues in seismic design and

performance evaluation is now well recognized by re- searchers as well as practicing engineers. The subject received special attention after the San Fernando

earthquake of 1971 when it became clear that damage of nonstructural elements not only can result in major economic loss, but also can pose real threat to

life safety. For example, an evaluation of various VA

hospitals following the San Fernando and other previ- ous earthquakes revealed that many facilities still structurally intact were no longer functional because of loss of essential equipment and supplies (Veterans Administration, 1976). Economic loss due to seismic nonstructural damage can also be considerable. A Ccise in point

is the seismic damage sustained by a seven-story hotel during the 1971 San Fernando earthquake. Of $143,000

in total damage in 1971-value dollars, only $2,000 was

structural damage while the remaining 98.6% wels nonstructural (Murphy, 1973). Recent federal estimates

of earthquake property damage after selected future major earthquakes in California are given in Table 1 (FEMA, 1981).

In practice, nonstructural elements are treated inde- pendently in the seismic design and evaluation of building structures. They are considered as being subjected

to forces applied by the structures to which they are at- tached. Today, major building codes and seismic design

guidelines exist which address the seismic force input

to various nonstructural elements.

Nationally, the Uniform Building Code (ICBO, 1988) is widely used for seismic design standards for structures and nonstructural elements, from which local jurisdic- tions and some federal agencies have developed similar seismic design requirements. The UBC formulates the design force for nonstructural systems as an equivalent

static lateral force applied to the approximate center of gravity of the system being analyzed. It is given by

Fp = ZICpW (1) where Fp is the lateral design force and Z is the seis- mic zone factor, taking values of 0, 3/16, 3/8, 3/4 and

1 for seismic zones 0, 1, 2, 3 and 4, respectively. / is the importance factor which takes the value of 1 for most structures, 1.25 for buildings with room occu- pancy loads of 300 or more and 1.5 for essential buildings such as fire stations and hospitals. W is the weight

of the nonstructural system and Cp is the lateral force

coefficient. A portion of the Cp-table stipulated in Title 24 of the California Administrative Code is reproduced

in Table 2 (State of California, 1985).

Some variations and refinements of the basic formula

(1) are found in other codes depending on whether

they consider earthquake response characteristics of the ground, attachment details, and vibrational chciracteristics of the structural-nonstructural systems. The Applied Technology Council (ATC), for example, has de- veloped some more elaborate guidelines. For seismic design of electrical and mechanical equipment, the for- mula is (ATC, 1978)

Fp = ACpMWpPm " " (2) where A is the effective peak ground acceleration, Cp

is the lateral force coefficient, M is the equipment amplification factor, Wp is the equipment weight, P is the performance level coefficient, and m is the attachment

amplification factor. The exception, it appears, is the nuclear power industry where much more stringent and rigorous standards

are recommended. For example, in seismic qualifica- tion cla^s IE equipment for nuclear power generation

stations, the recommended qualification methods consist of (IEEE, 1987):

(a) P''- licting the equipment's performance by analysi- including dynamic analysis of the combined

sti c cure-equipment system and nonlinear equipment response.

' National Center for Earthquake Engineering Research and Department of Civil Engineering, State University of New York at Buffalo, Buffalo, New York

14260

331
(b) Testing the equipment under simulated seismic conditions.

(c) Qualifying the equipment by the combination of

test and analysis.

(d) Qualifying the equipment through the use of experience data. Some of the provisions mentioned above have only re- cently been applied to newly constructed facilities and

for upgrading of precode nonstructural elements. Hence,

the Loma Prieta earthquake provided a good opportunity for assessing whether or not these new provisions

and standards represent a significant step forward in terms of improved seismic nonstructural performance.

For this purpose, rather than covering a broad spec- trum of nonstructural damage due to the Loma Pri- eta earthquaJce, this paper focuses on a selected list of nonstructural systems with emphasis placed on their performance evaluation with respect to the criteria under which they were designed, upgraded or constructed. These systems are (a) hospital facilities; (b) facades,

panels and glazing; and (c) pressure piping.

2. HOSPITAL FACILITIES

Following the 1971 San Fernando earthquake, in which

several modern hospital buildings and equipment sus- tained significant damage, the State of California gave

control to the State to develop hospital construction

staoidards and regulations, to assume responsibility for overseeing construction, aind to perform design plan

checking. The California Administrative Code, dis- cussed in Section 1, is the first to link geology, seis- mology, structural design and nonstructural system design. As a consequence, all hospital-related projects

have been subjected to review by the State and have

been made to conform to the requirements of the 1985

Triennial Edition of the State Building Code, Title 24, Part 2.

It is thus of special interest to observe seismic damage

to hospital facilities due to the Loma Prieta earthquake

in order to assess whether these State-imposed provi- sions have provided added margin of safety to these structures and their contents. A survey of 60 hospitals in the affected six-county area shows that a.11 hospital facilities performed well during

the event (Staehlin, 1990). Most, if not all, of the re- ported damage was limited to those facilities that had

not been upgraded to the above-mentioned provisions.

Minor damages to the nonstructural elements, however, were widespread. Numerous pieces of equipment

broke loose from their attachments and separated from

duct work or piping, causing some problems for the

affected facilities. Older vibration isolators and machine mounts in some cases literally exploded during

the event. Figures 1 and 2 show such instances at the Watsonville Community Hospital.

Elevators in these facilities posed a separate problem.

Of the 428 elevators in the survey, 2 . or 65.9% were

tripped by motion detectors and out f service for be- tween 10 minutes to three weeks, 50 of these sustained

damage ranging from loosened bolts attaching the guide

rails to twisted rails due to vibration of the counterweights.

3. FACADES. PANELS AND GLAZING

Preliminary observations indicate relatively good re- sponse of newer engineered buildings with corresponding minor damage to nonstructural elements such as facades, claddings and glazing. In downtown San Francisco, severe glazing damage occurred at a retail build- ing which is new in appearance. However, it was in fact constructed in 1946 as one of the first hermetically

sealed buildings in the city (EERI, 1989).

In older engineered buildings, some patterns of damage

to facades and claddings are apparent. Many build- ings showed loss of brick masonry facades and parapets. Figures 3 and 4 give an example of such damage

to a building in San Francisco's Mission District. In- deed, hundreds of multi-story masonry structures in San Francisco and elsewhere suffered extensive cracking and loosening of mortar. It is estimated that about

10,000-20,000 unreinforced masonry structures in San

Francisco sustained damage during the Loma Prieta earthquake.

Widespread damage to glass panels (Fig. 5), gypsum

board partitions, and suspended ceilings and floors (Fig.

6) was observed. This type of damage was common in steel buildings with masonry infill panels (A. Elnashai

et al, 1989). Such structures are typically more flex- ible than reinforced concrete structures and therefore impose large displacements on nonstructural elements.

As reported in (EERI, 1989), while nonstructural damage is considered relatively minor for each building in this event, the overall amounts to a large sum for area- wide damage. Some large corporations, such as Pacific

Bell and Pacific Gas and Electric, report nonstructural damage of up to $45 million or $50 million.

4. PRESSURE PIPING

The performance of steel piping at the Moss Landing

Power Plant of the Pacific Gas and Electric provided an opportunity to observe the effect of the Loma Pri- eta earthquake on many piping systems and their supports. The piping system at Moss Landing Wcis de- signed in accordance with the pressure piping industry's construction code (ASME ASA (now ANSI) B31.1

Code) in effect at the time of construction. The provi- sions of the Uniform Building Code was followed in the application of the lateral force criteria. The piping sys- tem was fabricated from carbon and low alloy (chromemolybdenum) steels and it is believed that such piping

systems should perform satisfactorily in an event of this intensity without loss of pressure retaining integrity, ex- cept where relative support motions exceed the cyclic strain limits of the materials.

Figures 7 and 8 illustrate some of the more important

observations at this site. Although pipe support defor- mations are evidently well into the plastic range, pipe

supports appear to have performed well with no evi- dence of loss of pressure retaining integrity or of damage to the presstire boundary. Figure 7 appears to sug332
gest that earthquake loads on pipe support elements might be allowed to induce "effective" stresses that

are well above current typical allowable stresses, i.e., perhaps some multiple of the material yield strength

(Haupt and Curtis, 1989).

These figures also show thermal saddles acting as bumpers between the support steel and the pressure boundary. These saddles are commonly used to spread the earthquake loads occurring from the pipe impact on

lateral supports over a larger area of the actual pres- sure boundary element. The saddles shown and other bumpers performed as expected.

Finally, several support elements broke entirely free from the structure to which they were attached due

to the imp2u:t of the pipe on the support. Where this occurred, however, it was subsequently found that the supports had been attached to the structure by tack welds (Haupt and Curtis, 1989).

5. CONCLUDING REMARKS

As in the past, nonstructural damage due to the Loma

Prieta earthquake is difficult to assess partly because

it is largely internal and invisible from the outside and

partly because of the lack of a systematic procedure for inventory and reporting of this type of damage. One

general observation that can be made is that nonstructural damage in this event is not as serious as originally feared due in part to the satisfactory performance of more recently engineered buildings. It is pointed out

in (EERI, 1989) that the Loma Prieta earthquake pro- duced ten to fifteen seconds of strong shaking with peak

horizontal ground accelerations ranging from 0.67g near

the epicenter to approximately 0.25g in the OaklandSan Francisco area. Had the ground motion duration

been significantly longer, a great deal more damage

would have resulted from this event.

As observed earlier, a significant amount of nonstructural damage was a result of support failure, which

could have been prevented had relatively simple protective procedures been instituted. Furthermore, observed

damages sustained by vibration isolators and machine

mounts underscore the continued existence of a serious

conflict between the need to anchor heavy machinery to the structure to avoid earthquake damage on the one hand, and the requirement that vibrating equipment

be isolated from the structure for vibration and ax:oustical control on the other. One possible solution to this problem may be the installation of braces which would

not touch the equipment base or piping system during

normal operation but become effective when excessive motions take place (Ayreb et al, 1973).

This work was supported in part by the National Center for Earthquake Engineering Research under Grant

No. NCEER-89-2002A.

7. REFERENCES

1. Ayres, J.M. et al, 1973, "Nonstructural Damage to Buildings," The Great Alaska Earthquake of 1964-' Engineering, National Academy of Sciences, Washington, D.C.

2. EERI, 1989, Loma Prieta Earthquake, October 17, 1990, Preliminary Reconnaissance Report 89-OS, El Cerrito, CA.

3. Elnashai, A. et al, 1989, The Loma Prieta Earth- quake of 17 October 1989: Seismological, Geotechnical and Structural Field Observations, ESEE Research Report No. 89/11, Imperial College of Sci- ence and Technology, London.

4. FEMA, 1981, An Assessment of the Consequences

and Preparations for a Catastrophic California Earthquake, Washington, D.C.

5. Haupt, R. and Curtis, D., 1989, Performance of Industrial Facilities, Pressure Piping Engineering

Associates, Inc., Foster City, CA.

6. IEEE, 1987, Recommended Practice for Seismic Qualification of Class IE Equipment for Nuclear Power

Generating Stations, IEEE, New York, NY.

7. International Conference of Building Officials, 1988,

Uniform Building Code, Whittier, CA.

8. Murphy, L. (Ed.), 1973, San Fernando, California Earthquake of February 9, 1971, Vol. lA, NOAA,

Washington, D.C.

9. Staehlin, W., 1990, A Report to the Building Safety

Board on the Performance of Hospital Buildings in the Loma Prieta Earthquake of October 17, 1989,

Sacramento, CA.

10. State of California, 1985, California Administrative

Code, Title 24 Building Standards, Sacramento, CA.

11. Veteran Administration, 1976, Study to Establish Seismic Protection Provisions for Furniture, Equipment and Supplies for VA Hospitals, Washington, D.C.

6. ACKNOWLEDGEMENT

Some of the photographs used in this paper were supplied by Professor Barry J. Goodno of the Georgia In- stitute of Technology, Mr. Ronald Haupt of the Pres- sure Piping Engineering Associates, Inc., and Mr. William Staehlin of the Office of Statewide Health Plain- ning and Development, the State of California. Their

generosity and assistance are greatly appreciated.

333
Table 1. Estimated Loss in Major Future Earthquakes

(Taken from FEMA, 1981)

Fault Loss to Loss ol Total loss buildings contents (S in billions) (S in billions) (S in billions)

Northern San Andreas 25 n 38 Hayward 29 15 44 Newport -Inglewood 45 24 69

Southern San Andreas 11 6 17

Estimates uncertain by a possible factor of two to three.

Table 2. Horizontal Force Factor Cp

(Taken from State of California, 1985,

Table No. 2-23J, Part B)

Category Direction of horizontal force

Value of C„

1. Mechanical and electrical components:

a. All mechanical and electrical equipment or machinery which is a part of the building service systems.

b. Tanks (plus contents) and support systems.

c. Emergency power supply systems in public school

buildings. Essential communication equipment and emergency power equipment in hospitals, such as engine generators,

battery racks and fuel tanks necessary for operation of such equipment.

d. Hospital piping, electrical conduit, cable trays and air handling ducts.

2. Hospital equipment when permanently attached to build- ing utility services such as: surgical, morgue and recovery room fixtures, radiology equipment, medical gas containers, food service fixtures, essential laboratory equipment, TV supports, etc.

3. a. Storage racks with the upper storage level more than S

feet in height (plus contents),

b. fHoor supported cabinets and bookstaclcs more than 5

feet in height (plus contents).

4. Wall hung cabinets and storage shelving (plus contents).

5. Suspended ceiling framing systems

6. a. Suspended light fixtures.

b. Surface mounted light fixtures.

Any direction

Any direction

Any direction

Any direction

Any direction

Any direction Any direction Any direction

Any direction

0.33

0.33

0.50

0.50

0.33

0.30

0.30

0.30

0.30

1.00

334
Fig. 4. Cars Crushed by Fallen Masonry,

Mission District, Saji Francisco

335
336
Socio-Economic Impacts of Lifeline Performance

Related to the Loma Prieta Earthquake

by

Maiyann T. Phipps* and Ronald T. Eguchi^

BSTRACT

\'

' feline performance during the October 17,

'89 Loma Prieta Earthquake was evaluated

. id documented in the paper entitled "Lifelines

prformance During The October 17, 1989

bma Prieta Earthquake" (Khater, et al., 1990).

pis companion paper examines the sociotonomic impacts of lifeHne performance,

lifelines considered are: transportation,

^ter, sewage, power, gas, and communication

stems. Among the most prominent

•cioeconomic impacts directly related to

eline performance are those resulting from

irthquake damage to the transportation

jtwork. The impacts on both individuals and

isinesses are discussed. As damage to other

'elines was limited, so too were the related

icioeconomic impacts. Observations related to

e impacts of each lifelines on the community

||id regional economy are presented.

I INTRODUCTION

ifelines, by definition, are critical systems

;sential for the well being of a community,

iften, the performance of a lifeline system is Measured not by the direct effects resulting

|tom system damage but by the indirect effects

lat result from its inabihty to function,

ritical services such as emergency water supply

)r fighting fires, electric power for business

id industry, communication services for

Coordination of response and recovery efforts,

ad transportation routes for import of needed

mergency supplies or evacuation are all ervices essential for the rapid recovery of a

fty after a major earthquake.

Economic and social losses as a result of

lifeline failures may indeed account for a large

percentage of the losses experienced after an

earthquake. Unfortunately, there is no

convenient way of quantifying or tracking such

losses. Some of the reasons for this difficulty are:

1. Lack of adequate measures for quantifying

social losses, i.e., inconvenience.

2. No systematic effort to collect data such as the number of homeless or relocations due

strictly to failures of lifelines.

3. Difficulty, in general, in quantifying

secondary losses, e.g., business interruption

losses.

4. Difficulty in categorizing social and

secondary economic losses by faciHty type,

i.e., building losses vs. lifeline losses.

5. Lack of systems models that account for the

interaction and interdependence between

different lifeline systems and businesses that

depend on lifeline services.

Fortunately, the resiliency of lifeline systems

and communities, in general, is such that in

^H.J. Degenkolb Associates, San Francisco, CA

94104

^Dames & Moore, Los Angeles, CA 90017

337
moderate earthquakes, many temporary repairs can be made to restore partial service to an

affected community. Because of this resiHency

the impact of Hfeline failures in recent U.S.

earthquakes has not been as catastrophic as it could be. In large earthquakes, such as a

repeat of the 1906 San Francisco earthquake,

socio-economic losses as a result of lifeline

failures would be expected to increase

significantly. Resources that are normally

available under mutual aid agreements may not

be available because these outside resources may also be affected by the earthquake.

Furthermore, earthquake damage will be more

severe, and in many cases, large portions of

lifeline systems may require complete

replacement.

It is important that future research efforts

record and attempt to quantify these

socio-economic losses. Performance models

should be developed that can provide even

order-of-magnitude estimates of these losses.

Breakdown of these losses by source, such as

lifeline failures, would also be important.

The following discussion summarizes, on a

quantitative basis, the socio-economic effects

caused by lifeline failures in the October 17,

1989 Loma Prieta earthquake. More work

needs to be conducted in order to collect

important data for measuring socio-economic

losses, and more research should be devoted to

developing socio-economic models.

2. SOCIO-ECONOMIC IMPACTS OF

LIFELINES PERFORMANCE

2.1 Transportation Systems

Of all the lifeline systems damaged by the

earthquake, none was damaged as severely and

dramatically as the transportation network.

Damage to highway structures was the most

significant cause of life loss and also carried

with it the greatest socio-economic impact on

the region. The collapse of the 1-880 Cypress

Street structure, the collapsed span of the

Oakland-Bay Bridge, damage to freeway

accesses leading into San Francisco, and

damage to Highway 17 in Santa Cruz forced a

temporary, yet overwhelming restructuring of

regional transportation systems and clearly

illustrated the region's economic dependence

on its transit system.

Physical losses to the highway system are estimated at $1 billion. The physical losses are only one indicator of the economic impacts on

the region, however. Economic losses due to business disruption and personal hardship also

accompanied the physical damage.

For the most part, alternate transportation

routes and/or transportation modes were

developed to bypass each of the devastated

highway segments. Consequently, major

societal impacts resulting from the damaged

highway system were averted. Socioeconomic

impacts were generally limited to modifications

to commute habits, longer transit times, and

business losses.

The socio-economic impacts related to the

collapsed Bay Bridge serve to demonstrate

some of the societal impacts. The Bay Bridge

collapse required that 250,000 commuters a day

seek alternate transportation to San Francisco

and/or develop alternate work patterns.

Because BART, the regional heavy rail system,

was essentially undamaged during the

earthquake, there was a transit system available

to accommodate commute traffic between San

Francisco and the East Bay. Most commuters

chose BART as an alternate. This shift in

commute patterns was effected by an

unprecedented cooperative effort among transit

agencies. Regional bus and train systems, ferry

service and BART formed a partnership to

assist the community at large. For example,

bus routes were modified to permit easy access

to BART stations and ferry terminals.

Representatives from all key agencies met

regularly to monitor, adjust and improve transit

338
service. Local newspapers and television

I maintained a key communication link between

j the transit agencies and the public.

Commute habits were substantially altered for

the short term. To accommodate the crowds

encountered on BART during peak commute

hours many riders altered their schedules to

permit early commuting. Companies

encouraged such changes by adopting "flex

I schedules". Some pre-earthquake commuters

shifted their offices to branches located near

their homes. The cooperative efforts of public

agencies and the flexibility of employers and

employees overcame the obstacle posed by the

loss of the Oakland-Bay Bridge. Consequently,

the San Francisco-East Bay link was maintained

and potential economic losses were minimized.

During several months following the reopening

of the Oakland-Bay Bridge, commute patterns

were observed to be somewhat of a blend of

pre- and post- earthquake patterns. Six months

after the earthquake, however, commute

patterns have generally returned to pre

earthquake days.

Some short term economic losses can be

directly or indirectly attributed in whole or in

part to the damaged highway transportation

system. Examples of the economic impacts,

focusing primarily on San Francisco, include the

following:

Decline in Retail Sales - The inability to drive

directly to San Francisco from East Bay

locations, crowding of the BART system, and

general fear invoked after the earthquake

caused people to postpone or redirect shopping

trips which were originally destined for San

Francisco. Both large and small San Francisco

retail enterprises suffered substantial revenue

declines immediately after the earthquake.

Because of the regional redistribution in retail

sales, large retail enterprises with branches

outside of San Francisco did not to suffer

appreciably. However, for those small

businesses without such branches the losses

were significant and unrecoverable.

Decline in Tourism - San Francisco's number 1 industry, tourism, suffered markedly after the

earthquake. Tourists were afraid to travel to

San Francisco fearing another earthquake

would hit and uncertain of the City's ability to

overcome the recent devastation. The hotel

industry - a key barometer of tourism - suffered

significant earthquake-related losses. In

October, San Francisco hotels lost nearly $32

million in revenues. Since October, convention

business has once again been restored.

"Discretionary" travel to San Francisco, which

accounts for half of all out-of-town tourism, has

dropped by as much as a third. (Examiner)

Decline in Restaurant Business - Restaurants

were forced to bear the brunt of decreases in

tourism as well as a decrease in the flow of Bay

Area residents into San Francisco. Many

residents of the region grew accustomed to not

coming to San Francisco during the month-long

Bay Bridge closing and hence, the earthquake

has had some short term as well as long term

impacts on the restaurant industry. In a twoweek period after the quake business in San

Francisco County declined 30 to 40 percent;

some restaurants lost up to 90% of their revenues. More than a month after the quake,

restaurant business on average was down 15%

to 25%. At least one restaurant closed

permanently. In Santa Cruz County one week

after the quake, business was down 60 percent.

In November and December, business was still down 60 percent. (Examiner)

Increase in Unemployment - Damage to

property coupled with the economic impacts on

tourism and related industry as described

herein led to substantial increase in

unemployment. In October, San Francisco's

unemployment claims were 50 percent higher

than the City's monthly average, while in

November the claims soared by 84 percent. In

Santa Cruz County where physical damaged to

339
property was the primary cause for

unemployment, the number of claims

quadrupled and in November they were twice

that of the county's average, (Examiner)

Lower Productivity - People missed work for

several days after the earthquake. Upon

returning, productivity was limited for many

reasons.

Some long term impacts of the damaged transit

system can also be identified.

Decline in Retail Sales One of the potential

long term impacts of the damage to the

highway system is demonstrated by the

earthquake's effects on merchants in

Chinatown. While the Bay Bridge was restored

to service one month after the earthquake, two

bridge approach ramps which direct riders to

Chinatown, Fisherman's Wharf and North

Beach, all popular San Francisco tourist spots, are still closed. These highway structures are of

construction similar to the 1880 Cypress

structure, suffered moderate structural damage

during the earthquake, and are hkely to remain

closed for a number of years.

Chinatown businesses are claiming that, six

months after the earthquake, business is still down approximately 25% from that prior to the

earthquake. They claim that this decline is due

primarily to the lack of direct access to the

area. Visitors driving to Chinatown must now

travel through many local streets, often in

bumper-to-bumper traffic. The Chinatown

businesses cannot support themselves with local

traffic only. Businesses rely on an infusion of

people from outside the City. An economic

study of the area is currently underway.

Enhanced Communication Among Regional

Transit Agencies Regional transit agencies

joined together after the earthquake to serve

the public need. Since that time, agencies are meeting regularly to discuss regional transit

issues and plan for the future.

Increased Spending on Seismic Safety The

State of California Department of

Transportation had requested state funds for

seismic strengthening of highway structures to mitigate the known seismic hazard posed by

numerous existing highway structures. In the

seventeen years prior to the earthquake, $54

million of state funds had been appropriated

for this purpose. Shortly after the earthquake,

state legislators appropriated $80 million for

seismic retrofit of state highway structures. The

funding was generated from a quarter percent

increase in the state sales tax. Additional

proposed legislation could lead to the

appropriation of up to $400 milhon to complete

the seismic strengthening of all known

hazardous highway structures (Roberts).

2.2 Electric Power Systems

Nearly one million PG&E customers lost

electrical power immediately after the

earthquake. Service was restored to 600,000

customers within 12 hours after the quake and

to another 300,000 within 36 hours. Physical

losses to PG&E totaled $1 million.

Because power outages were generally limited

in duration, the related socio-economic impacts

were likewise limited. In general, business

activities were suspended by most businesses

until full power was restored. Large businesses

and public agencies were equipped with

emergency generators to assist in recovery

efforts until power was fully restored. Some

companies obtained generators from outside

the area. Many small businesses simply waited

until commercial power was restored, incurring

the related business losses in the interim. The

loss of power contributed, in part, to local

unemployment.

Some businesses, which rely heavily on

commercial power for the manufacture of "hightech" goods lost substantial inventory as work in

progress was damaged. Those businesses

dependent on power for the preservation of

340
food or other items, lost inventory when power

was lost.

The interdependence of utilities was

dramatically demonstrated after the earthquake.

For example, the loss of commercial power

caused innumerable communications problems.

One common cause of telephone service

disruption was the lack of emergency power for

private branch exchanges (PBX). These PBX

systems are connected to outside telephone

lines but rely on in-house power. Where

reliable in-house emergency power was not

available, phone service to these systems was

lost for the duration of the commercial power

outage.

A similar interdependence of utilities was

demonstrated by some gas detection systems

which reportedly malfunctioned as a result of

the power outage. (Sherman)

The socio-economic impacts of power outages

after the Loma Prieta Earthquake appear to be

Hmited to the direct loss of foregone production

of goods and services. Long-term impacts

appear minimal. PG&E's customers will

generally fund the cost of repair, which in this

earthquake is relatively small and will not have

a dramatic impact on utility bills.

2.3 Natural Gas Transmission Systems

The Pacific Gas and Electric (PG&E) Company

service region covers most of northern and

central California and includes the entire

earthquake-affected area. PG&E receives the

majority of its gas through large diameter

transmission lines that enter California at the

Oregon and Arizona borders. There are gas

storage fields at various locations in the service

area; however, the two storage fields closest to

the earthquake epicenter were both well over

60 miles away.

The gas transmission lines and large-diameter

distribution mains experienced only three leaks

due to the earthquake. Unstable soil in the

Marina District of San Francisco caused

damage that resulted in the replacement of

approximately ten miles of cast-iron and steel

distribution mains with polyethylene plastic

pipe. Also in the Marina, 1,500 services to

5,400 individual meters were replaced in less

than five weeks. Three miles of distribution

mains were replaced in Los Gatos and

Watsonville. Other damage was scattered and

minor.

PG&E restored service to over 150,000

customers whose gas service had been turned

off. A total of 1,100 service personnel

participated in the relighting process, including

approximately 400 from other utilities in the

West. Within a week, service was restored to

all customers with undamaged piping.

Socio-economic losses as a result of gas

disruption were minimal in this event. Like

most of California, winters in the Bay area are generally mild; thus lack of heating fuel did

not pose a serious health threat. Lack of

natural gas does affect other daily functions,

e.g., cooking and bathing. In at least one

severely damaged area, i.e., the Marina District,

PG&E set up temporary bathing facilities for

residents until all repairs to water and natural

gas facilities had been completed.

Economic losses to industries requiring natural

gas as a fuel source are not known by the

authors at this time. However, gas service was

restored to most areas within several days.

Areas affected by longer outages, e.g., the

Marina District, Los Gatos, are largely

residential areas with relatively little industry,

Watsonville may be an exception.

2.4 Communications

Pacific Bell is the primary provider of local

telephone service in the Bay Area. For the

most part, damage to Pacific Bell facilities was

minor and did not affect service.

341
Overload of the system immediately after the

quake led to "slow dial tone". Pacific Bell

reduced the number of incoming calls to the area to provide maximum available service to

local residents. In most areas the impact of

phone service was to require waiting on the

order of a minute to receive dial tone.

For some businesses with Private Branch

Exchanges, power loss meant the loss of

their means of communication as discussed

previously. Some businesses had alternate

communication systems in place. However,

shortcomings in these systems were clearly

identified. For example, internal radio systems

were often rendered useless because of

overloading, i.e. too many companies were

using the same frequency. Some radio systems

were disabled when remote equipment was

damaged.

The interdependence of utilities was again

demonstrated by Pacific Bell's reliance on

commercial power. Pac Bell's central offices

generally have electric-generators to recharge

batteries which maintain phone service for

extended periods, without commercial power.

While only one central office was temporarily

disabled due to malfunctions of the back-up

energy supply, the need for a closer working

relationship between PG&E and Pacific Bell

was well-demonstrated. The need to

interconnect Emergency Operations Centers

with PG&E and Pacific Bell was also

demonstrated.

The short-term socioeconomic impacts of loss

of communications included the emotional

strain of being unable to reach loved ones to

determine their well being, loss of business revenue and delay in overall recovery efforts.

Every business, utility and public agency

impacted by the earthquake has gained an

appreciation for their dependence on

uninterrupted sources of power and

communications. Improved alternate

communications means are likely to develop.

2.5 Water and Sewage Facilities

In general, most of the damage to water supply

systems was to pipelines. Areas that were

severely affected included the Marina District

in San Francisco, the South of Market area in

San Francisco, parts of the East Bay (e.g.,

Oakland), cities in the South Bay (San Jose,

Cupertino, Campbell, Los Gatos and Los

Altos), and cities in the Monterey Bay area

(Watsonville, Pajaro, Hollister and Sunny

Mesa). Loss of electric power also affected the

operation of some pump facihties.

Water service was available to most

communities after the earthquake. The

greatest concern for communities affected by

pipeline breaks was contamination of water

suppplies. The following communities issued

advisories to boil water after the earthquake

(TCLEE, 1990):

o Los Gatos

o Watsonville o Hollister o Santa Cruz

o Scotts Valley

In general, no serious problems were

encountered with sewage and wastewater

systems. Power outages affected the operation

of some sewage treatment and pump stations,

and the loss of power resulted in some

untreated sewage being dumped into the San

Francisco Bay.

3. CONCLUSIONS

Lifelines are indespensible to the vitality and

viability of an economy. They are vital to life safety in the immediate aftermath of a

destructive earthquake and are necessary in the

subsequent reconstruction period. Direct

damage to lifelines also leads to secondary

342
effects including reduced business productivity,

loss of income, reduced employment, lower

consumer spending, and decreased property and

sales tax revenue. Each of these effects was

realized, to some extent, after the Loma Prieta

Earthquake.

Because the interruption of lifelines was in

general, limited in duration, the impacts that

lifeline damage caused by the Loma Prieta

Earthquake had on the Bay Area economy

were similarly limited. It is estimated, that in

aggregate, income loss over in the twelve

months following the earthquake is likely to

amount to less than one percent of the region's - total output. Unemployment directly resulting

from the earthquake is expected to amount to a

fraction of one percent of the region's total

employment of 3 million jobs (Munroe).

The only lifeline severely impacted by the

earthquake was the transportation network.

Both short and long term socioeconomic

impacts have been observed.

4. ACKNOWLEDGEMENTS

Partial funding for this limited investigation was

sponsored by the National Science Foundation

and the National Center for Earthquake

Engineering Research. This support is gratefully acknowledged.

5. REFERENCES

(1) Khater, M., Scawthorn, C, Isenberg, J., Lund, L., Larsen, T., and M. Shinozuka,

"Lifeline Performance During the

October 17, 1989 Loma Prieta

Earthquake," Proceedings of the 22nd

Joint Panel Meeting on Wing and

Seismic Effects, May 14-25, 1990.

(2) Munroe, Tapan, The Economic Impact

of the Loma Prieta Earthquake of 1989:

A Preliminary Look, Pacific Gas and

Electric Company, 1990.

(3) Roberts, James E., Caltrans, Manager,

Division of Structures, Personal

Communication, 1990.

(4) The San Francisco Examiner, January 14,

1990.

(5) Sherman, William J., Intel Seismic

Program Manager, Personal

Communication, 1990.

(6) Technical Council on Lifeline

Earthquake Engineering (TCLEE),

ASCE, Contribution to Earthquake

Engineering Research Institute (EERI)

Reconnaissance Report, 1990.

343
Lifelines Performance During the October 17, 1989

Loma Prieta Earthquake

by

M. Khater^, C. Scawthom^, J. Isenberg^, L. Lund-', T. Larsen^

and M. Shinozuka'^

ABSTRACT

Lifelines performance during the October 17, 1989

earthquake was evaluated and documented.

Lifelines considered are: transportation, water,

sewage, power, gas, and communications systems.

The Loma Prieta earthquake has demonstrated that earthquakes can still cause severe damage to modem

urban areas. Even though only few building were

actrually destroyed, major lifeline damage resulted,

demonstrating that the highway system can be

damaged, electric power can be lost to large areas, water mains can be severed, gas lines broken, sewer

lines ruptured and vital services lost. Among the most catastrophic seismic-induced events were the collapse of the elevated Cypress Street section of

Interstate 880 in Oakland, the collapse of a section of

the San Francisco-Oakland Bay Bridge, multiple

building collapses in San Francisco's Marina district, and the collapse of several structures in Santa Cruz

and other areas in the epicentral region. Damage

and business interruption losses were estimated as high as $ 6 billion. Human losses were severe.

KEYWORDS: Seismic damage; Lifelines performance; interaction.

Oakland Bay Bridge, multiple building collapses in San Francisco's Marina district, and the collapse of several structures in Santa Cruz and other areas in the epicentral region. Damage and business

interruption losses were estimated as high as $ 6

billion. Human losses were severe: 62 people died,

3,700 people were reported injured and over 12,000 were displaced. At least 18,000 homes were damaged, 960 were destroyed and over 2,500 other

buildings were damaged and 145 destroyed.

The Loma Prieta earthquake has demonstrated that earthquakes can still cause severe damage to modem

urban areas. Even though only few buildings were

actually destroyed, major lifeline damage resulted,

demonstrating that the highway system can be

damaged, electric power can be lost to large areas, water mains can be severed, gas lines broken, sewer

lines ruptured and vital services lost. In order to put

these effects into perspective, this paper records the performance of different lifelines (transportation,

water, sewage, power, gas, and communications

systems) during the Loma Prieta earthquake

.

1. INTRODUCTION

On October 17, 1989 at 5:04 p.m. local Hme, a Mg 7.1 earthquake occurred due to approximately 40 km

rupture along the San Andreas fault. The epicenter

of the 20 seconds earthquake was located near Loma

Prieta in the Santa Cruz Mountains about 16 km NE

of Santa Cruz, 30 km south of San Jose and about 100 km south of San Francisco ( Figure 1).

Major damage included the collapse of the elevated

Cypress Street section of Interstate 880 in Oakland,

the collapse of a section of the San Francisco1. EQE Engineering, 595 Market St.. San Francisco, CA

94105. .

2. Weidllnger Associates. 4410 El Camlno Real, Suite

110. Los Altos, CA 94022

3. Consulting Engineer, formerly Los Angeles Dept. Water and Power (ret.)

4. Department of CivU Engineering and Operation

Research, Princeton University, Princeton, NJ 08544

344
2. LIFELINES PERFORMANCE

2.1 Transportation Systems

A major effect of the Loma Prieta Earthquake was on

the transportation system, with the highway

network suffering the worst damage.

2.1.1 Highway

Several major highways and overpasses were damaged and rendered useless some for only a short

time, other for as much as a few years. Figure 2 shows the major highways and overpasses which

were closed due to earthquake damage.

The collapse of more than a mile of the Cypress

Street elevated section of 1-880 was the largest

contributor of deaths in the earthquake. The doubledeck Cypress Street overpass structure consisted of box girder decks supported by concrete frames

(Figure 3A). Support columns contained large

amounts of vertical reinforcing steel (typically #18)

which provided ample vertical load-resisting

capacity. However, only #4 bars, spaced at 12

inches, were used to tie the vertical steel and provide

shear resistance for the columns. These ties provided very little shear capacity, resulting in very sudden brittle failure of the overall beam /column

system (Figure 3B). Contributing to this failure were

the relatively large number of hinges in the frames,

typically reducing the structure to a deterministic

structure (i.e., redundancy). These hinges were

introduced due to concerns over shrinkage in the prestressed girders.

The 53-year old Bay Bridge (San Francisco-Oakland

Bridge) is the major surface link between the northern San Francisco Peninsula and the East Bay.

Approximately 300,000 commuters traverse the bridge each day. The loss of a 50-ft span of the upper deck of this bridge and subsequent damage to the lower deck caused major transportation

problems. The Bay Bridge is a complex structure composed of back-to-back suspension structures on

the west section, the world's largest single-bore

tunnel, and a cantilever truss bridge followed by a curved incline section to the east. During the earthquake the cantilever and incline sections

displaced longitudinally, resulting in the gap

increasing beyond the beam seat capacity of 5

inches, dropping the upper span onto the lower span

as shown in Figure 4.

Roads which were closed after the earthquake are given in Table 1 and shown in Figure 2.

2.1.2 BART

The Bay Area Rapid Transit District (BART) is a 71.5

mile, automated rapid transit system serving over 3

million people in the three BART counties of Alameda, Contra Costa, and San Francisco, as well

as riders from northern San Mateo County.

Due to the earthquake the transition structure located just west of the Oakland West Station was

damaged. The transition structure is a cast-in-place

reinforced concrete structure leading from the east end of the bay tube to the adjacent elevated

structure. The track grade is 3.98%. There are four

stringer beams, which line up with the rails. At each

bent, there are four concrete piles under the bent cap

beam. There are five spans, and each is forty feet long as shown in Figure 5. The stringer and bent cap beam for the second and fourth spans are cast monolithically with reirvforcement extending from

the top on the pile. This design provides a rigid

frame in the longitudinal direction for these spans. The third span is a floating span, and is supported

from the second and fourth spans by a bracket on

the side of the bent cap beam.

In the third span, which is the floating span, the ends

of all of the cap beam brackets were severely

damaged. At pier 4 (Fig. 5) south side (the sliding

joint) about half the width of the bracket has broken

out and fallen. The break extends back from the end

of the bent cap about four feet. On the north end,

there was a vertical crack which penetrated the full depth of the bracket and could seen on the bottom

for about 2 feet along the length. The third span was

temporarily supported by shoring on timber

cribbing. The damage to the transition structure was

most probably caused because the sliding joint did

not function as intended, and the reinforcing steel did not come all the way to the face of the bracket.

The transbay tube (Figure 6) performed very well during the earthquake. The profile was checked

immediately after the earthquake and little change was observed. The track alignment was excellent. The San Francisco side experienced a little movement, and the Oakland side moved about .75

in. longitudinally. The tube has a capacity of +/-

3.25 in. movement in the longitudinal direction, and

345
has a capacity of +6 in. movement m the vertical and

horizontal directions.

BART set its highest ever patronage records during

the closure of the San Francisco Oakland Bay Bridge.

Weekday trips averaged 347,122, compared to 257,500 trips after the bridge opened, or an increase

of approximately 40%.

2.1.3 Port Facilities

This Section summarizes the damage to the marine

facilities of the Port of Oakland. There are five ports

in the Bay Area (Oakland, San Francisco, Redwood

City, Richmond, and San Jose) but only Oakland

sustained significant damage. Initially after the earthquake, total damage to the Port of Oakland was

estimated to be on the order of $100 million, with

approximately 60% of that corresponding to damage

sustained to the airport (the air facilities portion of

the Port of Oakland). In an interview with port

officials in March, 1990, this estimate has been

revised down to between $50 and 60 million, with

the airport damage approximately $20 million of the

total. Thus, the total damage to the marine facilities

is estimated to be between 30 and 40 million dollars. This figure is primarily for the repair and/or the replacement of facilities and infrastructure components. The cargo handling components are primarily owned and maintained by the tenants.

The peak ground acceleration at this site is assumed

to have been on the order of 20% of gravity. The

collapsed portion of the Cypress viaduct is within

1/2 mile of a considerable portion of the marine

facility.

Damage to the Oakland marine facilities is most

easily examined by looking at the damage to the components. The cargo handling operations

sustained minimal damage, emerging from the earthquake substantially operational. The facilities sustained a significant amount of damage, with

temporary fixes necessary in some areas. The

infrastructure was damaged to a degree, but was

fundamentally functional soon after the earthquake.

CARGO HANDLING OPERATIONS: The cargo handling operations consist of items such as liquid

storage tanks, bulk storage elements, container

storage elements, cranes, etc. The cargo handling

operations sustained minimal damage, emerging

from the earthquake substantially operational.

Estimated damage to cranes is 0.5% of their value,

and the estimated damage to the remainder of the cargo handling operations is negligible.

Total damage to the 23 cranes is estimated to be on

the order of $1 million. Assuming a value of $8

million per crane ($184 million total), this corresponds to a damage ratio of approximately

0.5%. Damage to the cranes was limited to minor

racking and pounding, with no incidence of loss of

stability. Not all of them are 100% functional

presently, but that is due to the settled condition of

the dike, and not to crane damage. The rails upon

which these cranes are supported are supported on

piles. Differential settlement between the rails varied from zero to 2 feet. It is estimated that these cranes could operate with a differential settlement of

8 inches.

Cargo containers in the facility did not overturn. Container storage in the marine facility consist of both containers on chassis and stacked containers. Stacked containers were both in grouped stacks, and

free standing, with no adjacent stacks. Stacked

trailer chassis did overturn, but were not damaged.

There are no major liquid or bulk storage elements

within the marine facilities. Much of the bulk cargo

is break bulk. This cargo is not known to have

sustained much damage. It is the opinion of the Port

that damage to goods in transit was minimal.

FACILITIES: The facilities at the Port of Oakland

consist of items such as buildings, berthing

structures, retaining structures, hydraulic fill, etc. The facilities sustained a significant amount of damage, with temporary repairs necessary in some

areas to allow operations to continue. Damage to buildings varied, but was very minimal. The

berthing structures underwent the greatest damage,

attributed to the use of batter piles. Settlement of the hydraulic fill in the port area disrupted the use of the

facility.

Damage to buildings varied There is an estimated

$1 million damage to the office structure in the APL

Terminal (berths 60-63). This is a reinforced concrete

346
frame structure, with damage primarily limited to failure of the unreinforced masonry infill in the

stairwells. The container freight station at berth 10 had minimal damage, with some broken studs in the

walls, and minor spalling of the columns. One

container freight station is slated for demolition, but

this is primarily due to the need to rebuild the wharf

structure underneath it (Howard Terminal, Berth

69). The remainder of damage to buildings is minimal.

Damage to the berthing structures is substantial, and

the cost to repair these facilities is a significant

portion of the total repair costs. Damage is primarily

due to the use of batter piles upon the wharfs. The

batter piles were inadequately connected to the wharf structure, and this connection was typically

heavily damaged, with very little or no residual

strength. Differential settlements, both laterally and

vertically, leave the wharfs pulled away from the dike in many places from 2 to 8 inches. Wharfs with

all vertical piles performed considerably better, with

very minor cracking of the piles. Piles are primarily

concrete, with a few concrete jacketed timber piles.

New construction is slated to utilize all vertical piles.

All of the wharf structures have been surveyed since

the earthquake, with damage tabulated, including a map of all the piles, and the damage sustained. One

wharf structure (Berth 26) sustained damage at the

interface where the structural system changed from

batter piles to an all vertical system.

The use of retaining walls is not very prevalent in the Port of Oakland. There was one cracked

concrete retaining wall at berth 69 (Howard

Terminal). All wharfs are on individual piles and

there are no sheet piles used at this facility.

The settlement of the hydraulic fill varied. By far the greatest settlement was near the end of the Seventh

Street Terminal, where the fill settled up to 5 feet. This terminal is directly above the transbay BART

tube, which may have accounted for some of the settlement. The Seventh Street Terminal was

perhaps the most damaged portion of the entire

facility, with many other extensive vertical displacements ranging up to 2 feet, and lateral displacements ranging up to 1 foot. Settlements in some areas preclude the continued usage of the large

container cranes. The effects of ground settlement and liquefaction were present generally throughout

the port. These effects included sand boils, settlements, and large cracking of the pavements.

INFRASTRUCTURE: The infrastructure consists of roadway and rail transportation (tracks, pavements,

tunnels, bridges, causeways, signals, etc.); utilities (power, communication, water, natural gas, etc.; pipelines, etc. The infrastructure was damaged to a degree, but was fundamentally functional soon after the earthquake. Roadways were mostly

undamaged, and only a small portion of the railways were unusable, mostly due to excessive ground

settlement. Underground water pipes broke, but

there is no indication of any major damage to other

underground pipes. Emergency services were

mostly untested, there were no fires or life- threatening collapses.

By and large, the roadways in the Port area were undamaged. The roads are used for extra heavy

loads, with allowable axle loads greater than on

most roads. This extra strength may have been a

factor in their satisfactory performance.

Rail facilities underwent varying amounts of damage. At the Bay Bridge Terminal (Berth 9), railroad tracks buckled both horizontally and

vertically. At the Seventh Street Terminal, rails going onto the wharf structure were damaged from

differential settlement of the wharf and the piles

supporting the rails. Damage at the two intermodal

yards (Southern Pacific and Union Pacific) was

reported as being minor.

One 18" water main burst, and was eventually shut

off (Berth 20). Upon the resumption of service, several more (estimated at about 20-30) leaks were

discovered. The capacity of the water supply after the earthquake remained mostly untested. In the event of a fire, they would have had to leave the burst water pipe in the circuit, having to deal with a large loss of pressure and water volume.

The facility was without power after the earthquake,

due to damage distant from the facility. Power was

restored without any problems.

There were no reported fuel leaks. The facility does

not have any oil storage tank or underground fuel pipes. Ship fueling is performed via barges from

Richmond.

347
2.2 Electric Power Sys tems

The electric power systems in the earthquakeaffected ares are operated by the Pacific Gas and

Electric Company (PG&E). Figure 7 shows the

electric power facilities and the transmission routes

in the earthquake-affected area. About 1.4 million customers lost electric service in the minutes after the earthquake. Most of them were in San Francisco,

San Mateo, Santa Clara, Santa Cruz, San Benito,

AJameda and Contra Costa counties. Within 48

hours, service to all but 26,000 customers was

restored. In San Francisco, a 217-megawatt unit at Potrero Power Plant and two 106-megawatt units at the Hunter's Point Power Plant tripped off line when

the quake hit. The U.S. Navy loaned PG&E the services of the frigate USS Lang to provide

emergency steam to heat the feedwater for the Potrero Unit. The two Hunters Point Units that tripped off the line were returned to service as the

electric load began to climb on October 19. In San

Jose and San Mateo, two key substations (Metcalf

and San Mateo substations) suffered major damage,

contributing to the interruption of electric service in San Francisco and the area south of San Jose. At

Metcalf substation in south San Jose, heavy damage

occurred in the 500,000-volt switchyard The San

Mateo substation is an important link in the system

that delivers bulk power into San Francisco. Damage to a transformer sharply reduced the amount of power that could move north. At Moss

Landing Power Plant on Monterey Bay 30 miles

south of the quake epicenter, a 750-megawatt

generating unit was damaged and went down, as were the high-voltage switchyards connecting the plant to the transmission system, the Diablo Canyon

Nuclear Power Plant near San Luis Obispo, about

140 miles south of the epicenter, had no damage and

continued to operate without any interruption.

Emergency repairs reconnected the transmission

links serving San Jose and San Francisco and new

substation equipment was flown in with the help of

the U S. Air Forces.

2.3 Natrural Gas Transmission Systems

Two major natural gas transmission routes feed the San Francisco Bay Area system, with some systems

dating back to the 1940s. From the south through

the Santa Clara Valley, supply is via two large

diameter pipes; and from the east, three lines are used for supply. Other transmission lines are used

to feed the outlying areas.

There were only three failures reported to gas transmission and large distribution lines. There were leaks in a 20-in. semi-high-pressure welded

steel distribution line in Oakland, a 12-in. line in Hollister, and an 8-in. line in Santa Cruz. In the sand v fill soil of San Francisco's Marina District, gas

lines were subjected to greater stresses and were

severely damaged. Damage was distributed throughout the city with more than 400 breaks in mains, service, and meter locations. A construction

force of over 300 workers worked to rebuild about 10 miles of gas hne at an estimated cost of S20 million,

which was completed in about three months.

In the epicential area in the Santa Cruz \iountains,

distribution system damage was restricted to three small line leaks. Two of these occurred at locations where gas lines were attached to bridge structures

that suffered damage. A third leak occurred in a section of old line constructed in 1948. This line was

in the process of being replaced. Los Gatos,

Watsonville, Hollister, Richmond, San Jose, Oakland

and Alameda also reported damage to either mains

or services.

In the hours after the quake, about 153,000 customers shut off gas service to their homes or

businesses. Where customers shut off their gas

service, a PG&E service person checked for possible

leaks, re-opened the service line, checked again for any problems, then re-lit pilot lights where

necessary. PG&E service personnel used gas

"sniffing" devices to check for leaks from natural gas

lines. If escaping gas was noted, lines were

uncovered and either replaced or repaired. In the hard-hit areas, inspectors checked for pockets of gas

before allowing re-occupancy, and before electric service could be restored.

Gas service was essentially back to normal

throughout the system in four days except for San

Francisco's Marina District and hard hit areas in the Santa Cruz, Watsonville, Los Gatos area. The heavy

damage to the gas system in downtown Los Gatos was repaired within two weeks after the quake.

2.4 Communications

Pacific Bell is the primary provider of local telephone service in the affected area. The damage

to Pacific Bell facilities was minor and did not affect service. In general, telephone systems performed

348
better than expected, especially considering the seismic forces that equipment in the epicentral area had to withstand. For example, in Watsonville a four-story reinforced concrete shear-wall structure

built in the 1950's Wcis subjected to peak ground

acceleration of 0.39g in the horizontal direction and

0.66g in the vertical direction. The peak horizontal

acceleration on the roof was 1.24g. The building

contains three floors of telephone switching

equipment. In addition, it has extensive piping,

cable trays, and heating and ventilating systems.

During the earthquake, when off-site power was

lost, the emergency diesel generator started up

immediately and it and the batteries in the building

kept the system on-line through the major

aftershocks.

The loss of commercial power caused some

communication problems. Telephone systems are designed to run on batteries for several hours. Central offices usually have electric-generators to recharge the batteries so that phone service can be

maintained for extended periods without

commercial power. There were some problems with

electric-generators and in one case service at an

office had to be reduced until additional backup

power could be obtained.

A more common cause of telephone service disruption was the lack of emergency power for private branch exchanges (PBXs) which are privately

ov^ed phone systems used by companies and large

organizations for telephone service v^rithin a facility. These connect to outside lines provided by the local telephone company and use building power when

this was lost, so was telephone service.

2.5 Water and Sewage Facilities

Water supply for the Bay Area is primarily imported

via long aqueducts from the Sierra Nevada or the Sacramento Delta. A small p)ortion of the supply,

especially for Santa Cruz and Monterey Counties is pumf)ed from local groundwater basins or collected

as local surface water from steams and springs.

Wastewater treatment facilities for most of the area

are both secondary and tertiary treatment facilities and discharge treated water into the San Francisco

Bay, Pacific Ocean, Monterey Bay, or sewage

lagoons.

A preliminary survey of earthquake damage was

conducted by members of the American Society of

Civil Engineers, Technical Council on Lifeline Earthquake Engineering, Earthquake Investigation

Committee, who contacted 25 water agencies and 12 sewer entities, only a portion of the total number of agencies in the area.

The most significant impact on water and sewage

lifeline systems of the Loma Prieta Earthquake was

the loss of commercial power for a long period of

time. Those agendes without emergency power

were without power from a few hours to several days. Commercial power, which is supplied by one

utility, had significant damage to several high

voltage substations which caused most of system in the earthquake area to lose power. The power

supply was restored carefully and slowly to service areas and only after checking for gas leaks to avoid

the potential for fire. Some of the emergency power

supplies had problems because of limited fuel capacity or failed to op»erate because they had not been tested.

There was no reported damage to any of the four

major aqueducts or any of the aqueduct storage

reservoirs supplying the area. There was minor

damage to a few local reservoirs consisting of earth embankment and abutment cracks and concrete spillway cracks. Repairs were made by excavation

and recompaction and epoxy grouting. All the reservoirs were at a low level at the time of the event because of the third-year drought conditions in California and all have been authorized for filling.

There were approximately 800 to 1,000 water main

breaks due to the earthquake primarily in the unstable areas along the shore of San Francisco Bay

and along the rivers and estuaries entering Monterey

Bay. Most of these areas suffered from liquefaction. In general, breaks were longitudinal and

circumferential and occurred in small diameter cast iron (sand cast) with fixed (cement or lead caulked)

bell and spigot joints pipe; and were due to tension,

compression, and bending due to differential settlement. Corrosion of pipe and fittings also played a part in some of these breaks.

In the San Francisco Marina District, the 4-, 6-, and 8- inch mains were installed in the 1930's and were

replaced following the earthquake with ductile iron pipe with rubber gasket joints. It is estimated that

349
TOO of the 150 breaks in the San Francisco system

occurred in the Marina area. A majority of the main

breaks in Alameda occurred on 4- and 6-inch

diameter welded steel pipe and were due to corrosion at the joints. Leaks occurred in the San

Jose-Los Gatos area on mains ranging in size from 4-

to 37-mches of steel, cast iron, asbestos cement, and

ductile iron; three-quarters of which were in 25- to 30-year old steel pipe with limited corrosion

protection. The types of leaks were holes in the pipe

due to corrosion, pulled joints and circular breaks. More than one-half of the leaks were due to holes in the pipe due to corrosion.

In the Monterey Bay area damage to water mains

also occurred to mains in unstable areas due to liquefaction. Failure modes included bell cracking,

circumferential, longitudinal cracking and push-on

joint failures. A majority of the breaks occurred in cast iron pipe with lead caulked joints and small

diameter galvanized iron pipe at the threads It was

reported that ductile iron pipe and asbestos cement

pipe with rubber gasket jomts performed well. It is not known how much of these kinds of pipe is in the older areas subjected to liquefaction.

Service connection breaks occurred in galvanized

iron, copper, and polyvinyl chloride pipe.

Relatively few sewer breaks have been identified,

because most sewers are gravity flow and water does not appear at the surface of the ground unless

there is a blockage. It may be assumed, however,

there are at least as many sewer main breaks as water breaks, especially in the same areas of unstable ground. Contacts with sewage agencies

indicate they are in the process of performing

television (video camera) surveys of the mains to determine the extent of the damage. Breaks in sewer

pressure lines (force mains) were identified when

power was restored to the sewage pumping stations.

There was little damage reported to pumping

stations and pumping wells, except for outages due

to the loss of commercial power. Three wells constructed without gravel pack became sanded

near Santa Cruz and have been redrilled and another

well in Watsonville had a broken discharge pipe.

There was significant damage to small (10,000 to 100,000 gallons) bolted steel, welded steel and

redwood tanks, especially in the Santa Cruz

Mountains. The damage consisted of collapses, shell buckling (elephant foot), roof buckling and

inlet/outlet problems of tanks which had little or no

seismic considerations in the design. A 1.1 million

gallon post-tensioned concrete tank failed at the

vertical joint. The tank was constructed of precast

vertical wall panels wrapped with post-tensioned

wire and coated with gunite. There was evidence of corrosion which may have contributed to the failure.

In two iriStances, undamaged water storage tanks were drained. In Hollister, an unreinforced masonry

building collapsed onto an adjacent building which

had a wet (fire) standpipe. The debris from the collapsed building pulled a coupling on a 6-inch

private fire service releasing an uncontrolled flow of water which almost drained a 2 million gallon tank

before it could be controlled.

In San Frandsco, the Auxiliary Water Supply System

(AWSS) provides high pressure supply for emergency fire protection. This system relies on

local reservoirs and tanks for supply and is completely separate from the municipal system. A

water main on this separate system broke several miles away from a 750,000 gallon tank, which led to temporary loss of pressure in the system (see Scawthorn and Blackburn, 1990). These two events

indicate the need for monitoring and controlling the outflow from the storage system after a disaster such

as an earthquake.

Damage occurred to a number of water and

wastewater treatment plants due to wave action

(sloshing) in the South San Francisco Bay area. A

laboratory person in Los Gatos reported the wave

action almost reached the top of a 30-inch protective

railing around a clarifier basin. The damage at an 80

mgd plant was due to the differential movement

between the up-flow center feed clarifier and the rectangular walls of the basin. The launderers

(wares) pulled away from the wall or pounded

against the wall. Some of the launderers fell to the bottom and jammed the sludge scrapers. At a 40 mgd wastewater treatment plant wave action caused

fiber glass scum troughs to be damaged in rectangular clarifier basins. The troughs were

attached to the concrete wall with a fabricated fiber glass bracket that broke at the angle point. The

troughs fell to the bottom and jammed the sludge

sweeping scraper. Both plants continued operations

350
by placing basins not in use back in service.

The major lessons learned from the Loma Prieta earthquake were to continue to use flexible joints in water and sewage piping, provide emergency power

for essential facilities, monitor the storage level in tanks and reservoirs to control outlet flow and to consider wave action in the design of treatment

plants.

3. OBSERVATIONS OF PIPELINE RESPONSE

AT PARKFIELD, CALIFORNIA

An experiment to measure pipeline response to travelling wave effects and to lateral offsets is located on the San Andreas Fault at Owens' Pasture near Parkfield, California approximately 175 km to the south of the epicenter of the October 17, 1989

Loma Prieta Earthquake. A magnitude 6 earthquake

is predicted by the U.S.G.S. to occur here within the next few years. The experiment involves an array of welded steel and ductile iron pipe segments which

are placed across a strand of the fault. Instrumentation includes strong-motion

accelerographs, strain gages on the welded steel pipes, and relative displacement transducers at the

joints of the ductile iron pipes. The threshold trigger

levels for recording were set at 2 gal. to record

foreshocks of the predicted Parkfield-Cholame

earthquake and other low amplitude seismic events. Usefvd records of pipeline strain and ground

motions were obtained from the Loma Prieta earthquake, which caused 20 gal. peak accelerations

at Owens' Pasture.

3.1 Observed Response in Loma Prieta Earthquake

Peak accelerations of about 20 gal. were recorded in both surface (0-40 sec) and body wave (about 40-80

sec) phases. A triangular region of ground defined

by the three seismographs is assumed to deform

with constant strain. When the measured horizontal

components of ground displacements are imposed

on vertices of a constant strain finite element, peak

transient values of ground strain of order .001 are found.

Time histories of strains were recorded on the welded steel pipe segments beginning in the interval 34-40 seconds after the start of the ground motion

records, which corresponds to the surface wave

phase. It is possible that the body wave phase did

not induce pipe strains high enough (20 microstrain)

to trigger the memory of the data logger; it is also possible that peak strains exceeded 20 microstrain

during the body wave phase but at a frequency too high to be resolved by the data logger whose upper

limit is currently about 1 hz.

The maximum strains in the tension pipe segment

are tensile because transient compressive and tensile strains are superposed onto permanent tensile strain induced by right lateral strike slip; transient compressive and tensile strains in the compression

segment are superposed onto permanent

compressive strain. The amount of offset that would

cause the observed permanent pipe strains is about

1.3 mm. Coseismic creep derived from the U.S.G.S. creepmeter XPKl is about 1.8 mm. Apparently, some slip occurred at the pipe-soil interface.

The data also illustrate the following points:

1. The pipeline strains on opposite

sides of the springlines are approximately equal. This

shows that, at least in the horizontal plane, direct extension or compression

dominates and bending is less significant.

2. The frequency content of the

direct ground strain and of the pipe strain time histories agree

closely. In the interval between

34 and 50 seconds, the ground

and pipe strains at a number of

stations are nearly the same

with peak and valleys of the same amplitude occurring

almost simultaneously.

3. The shifts in the amplitudes of

the pipe strain time history

records correspond to the development of a permanent

ground strain that appears to occur between 40 and 50 seconds. It is apparently

associated with lateral offset as measured in the creepmeter

XPKl.

351
4. CASE STUDY O N SEISMIC PERFORMANCE

OF LIFEL INE SYSTEMS

As mentioned above, lifeline systems serving the Bay Area performed reasonably well under the Loma Prieta Earthquake, even though isolated

failure were significant. It is anticipated, however,

that, given time, earthquakes with larger magnitudes

will occur, possibly at a more unfavorable location in the Bay area Therefore, a case study has just been

initiated under the sp>onsorship of the National Center for Earthquake Engineering Research

(NCEER). The case study purpose is to investigate

how lifeline systems such as water, power and gas

delivery systems, telecommunication and

transportation systems serving a community

performed under the Loma Prieta Earthquake. A

community will be chosen for the case study most

probably among Los Gatos, Watsonville, or Santa Cruz, taking into consideration the extent of damage, size of community in terms of geography

and population, number of lifeline systems involved,

degree of utility and governmental cooperation, etc.

The following NCEER researchers are participating

in this project: A.H-S. Ang (University of California,

Irvine), T. Ariman (University of Tulsa), B, Jones

(Cornell), J. Isenberg (Weidlinger Associates), M.

Khater (EQE Inc.), F. Kozin (Polytechnic University

of New York), M. O'Rourke (RPI), T. O'Rourke

(Cornell), M. Phipps (H. J. Degenkolb Associates)

and C. Scawthorn (EQE Inc ) (Table 2). There are six specific tasks, some of which can be performed

simultaneously. First, physical and operational

information on each system will be gathered and

documented. Second, on the basis of the information thus gathered and documented,

physical and operational models will be developed

for the system. Using these models, the state of damage and that of unserviceability will be

simulated for each system under the Loma Prieta Earthquake and compared with the actual states of damage and unserviceability for model verification. Third, performance interaction among lifeline systems will be reconstructed on the basis of available data and also modeled. Fourth, the nature and extent of seismic preparedness on the part of governments, utilities, industry, and citizens will be

examined as well as to what extent such

preparedness, if implemented, indeed mitigated the seismic disaster. Fifth, an analytical model recently

developed by Kozin for resource allocation to implement repair and restoration of damaged

lifeline systems will be tested against what actually

happened, and revised if necessary. Finally, the

societal impact of simultaneous and multiple lifeline system failures will also be investigated. Each of these tasks are listed in the attached table together

with the names of investigators and brief comments

on the task contents.

It is believed this case study will provide valuable

insight into the physical, operational, governmental,

and societal consequences of lifeline system failure caused by earthquakes for the community as a whole. Lastly, study of the system interaction issue

is also of substantial interest from both the technical and societal points of view.

5. REFERENCES

C. Scawthorn and F. Blackburn. Performance of the San Francisco Auxiliary and Portable Water Supply

Systems in the 17 October 1989 Loma Prieta Earthquake.

Fourth U S. National Conference on Earthquake

Engineering. May 20-24, 1990. Palm Springs,

California.

6. ACKNOWLEDGMENT

We are indebted to Dr. Jim Gates and Mr Jerome

Houke from Caltrans, Mr. Sy Mouber and Mr. W.

Belding from BART, Dr. Edward Matsuda and Mr. William Rowe from PG&E, and Mr. Gerald Serventi

and Mr. Thomas Daniels from the port of Oakland

for their assistance in documenting the performance

of lifelines during the Loma Prieta earthquake.

Support from the National Center for Earthquake

Engineering Research, SUNY-Buffalo, is also gratefully appreciated.

352
Table 1 MAJOR ROADS CLOSED AFTER THE

EARTHQUAKE

Martinez-Benicia Bridge "closed to trucks in both directions

1-80 and 1-880 closed between Berkeley and I- 980 in Oakland due to the collapsed Cypress

section

Bay Bridge closed because of structural damage

Embarcadero Freeway closed due to earthquake-induced damage

Interstate 280 closed from U.S. IGlto

downtown San Francisco

U.S. 101 northbound at Highway 92 overpass dosed due to debris

Highway 9 dosed at the San Lorenzo River

Bridge (near Santa Cruz) due to structural damage

Highway 17 dosed from Scotts Valley to Highway 9 due to landslides

Highway 1 north dosed at Struve Slough

Bridge due to a collapsed bridge

Highway 129 closed at Aromas Road

Highway 25 closed from U.S. 101 to 15 miles

south of HoUister due to rockslides
Figure 1: Epicenter of the Loma Prieta Earthquake

354
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Street overpass structure

357
SOURCE COaECe OFENGINEEaiNG UNIVERSfTY Of OLIFOBNIA *' 8£flKE.Ey

Figure 3B: Poorly detailed and inadequate horizontal ties were primarily responsible for the Cypress section column

fai 1 ures

358
t SAN FRANCISCO OAKLAND

SAN FRANCISCO OAKLAND

SOURCE: COLLEGE OF EhJaiNEERlNQ. UNIVERSITY OF CAUFOBNIA AT BERKELEY

Failure of the Bay Bridge during the Loma Prieta

Earthquake

359
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plan and profile

360
Figure 6: Plan and profile of trans-bay tube

361

Summary Report on the Loma Prieta Earthquake of

October 17, 1989

by

Toshlo IWASAKI^'

ABSTRACT

I

I On October 17. 1989, the Loma Prieta

Earthquake struck the Santa Cruz region.

California. U.S.A. The Japanese government

sent a study team for the purpose of investigating characteristics of damages caused by

the earthquake. This paper presents a summary of the full report prepared by the

study team, and emphases on the outline of

the damage, strong motion observation,

liquefaction, slope failures, damage to

lifeline facilities, measures against disaster,

voluntary activities by the residents and

lessons learned from the earthquai<e.

KEYWORDS : Loma Prieta Earthquake,

Earthquake Disaster,

Strong Motion Observation,

Liquefaction. Slope Failure,

Lifeline Facilities,

Measures against Disaster

1. INTRODUCTION - . ,

The Loma Prieta Earthquake with - magnitude of 7.1 occurred in the Santa Cruz

region. California. U.S.A.. at 5:04 p.m. ( Pacific Daylight Saving Time ) on October 17 ( Tuesday ), 1989. The earthquake struck not only the epicentral region but also San

Francisco Bay area located about 90km north

of the epicenter. As a result of the

earthquake, 62 persons lost their lives and

important public facilities and buildings were

heavily damaged.

The Japanese government sent a study

team consisting of 16 experts of the

concerned organizations and institutions in order to contribute to development of the measures against earthquake disaster not only in Japan but also in world-wide

countries where occurrences of earthquakes

are great threat. The Investigation started

from mid-November for two weeks Just after

finishing emergency rescue activities for

suffered people. An Investigation on the

damages in the epicentral region and San

Francisco Bay area was made with a great

assistance and support of many organizations

concerned with disaster prevention including

U.S. federal organizations, the California state government, local ( county and city ) governments and private firms. The Investigation by the Japanese team was successfully completed with a great deal

of supports by Dr. R.N. Wright, the U.S. Chairman of the Panel on Wind and Seismic

Effects, UJNR. Mr.N.J.Raufaste,

Secretary-General of the Panel, and

Dr. H.S.Lew, Dr. A. G. Brady, and Mr. J. H.Gates ( members of the Panel ). The original full report ( in Japanese,

about 300 pages ) was published in January

1990. The contents of the report are as follows:

Preface

1. Introduction

2. Natural Conditions and Social Conditions

3. Systems of Earthquake Prediction and

Precursors of the Loma Prieta Earthquake

4. Outline of the Event, Crustal Movements

and Strong Motion Observations

5. Soil Liquefaction and Slope Failures

6. Damages to Roads and Bridges

7. Damages to Buildings

8. Damages to Lifeline Facilities

( Except Roads and Bridges ) 9. Disaster Preparedness Administrations

and Emergency Measures

10. Public Responses ( Voluntary Activities

and Contributions of Volunteers, and

Public Education ) 11. Closing Remarks ( Summary of Findings.

Discussions with U.S. Experts, Proposals,

and Acknowledgments )

This paper presents a summary of the

report with an emphasis on the outline of

the damages, ground failure, strong motion

data, damages to lifeline facilities ( road

facilities and others ), measures for disaster

prevention and voluntary activities by the

residents and conclusions ( lessons learned

from the earthquake ).

2. OUTLINE OF DA.MAGES

Fig.l shows the epicentral region and San

Francisco Bay area affected by the Loma

Prieta Earthquake. Major damages were developed in the region of Watsonvllle located

south of the epicenter to Martinez located

north of Richmond.

FEMA ( U.S. Federal Emergency Management

Agency ) and OES ( Office of Emergency

Service, State of California ) reported an estimation of damages as DNumber of death : 62 persons

( Besides a few lives were lost by Indirect

effects such as heart attack )

Director General, Public Works Research Insti- tute, Ministry of Construction

363
2) Number of the Injured : 3.208 persons

3) Total cost of damages : estimated as 70 to 80 billion dollars

( An official estimate by the Federal or State governments has not yet released ) 4) Suffered people : about 55,000 persons

3. OUTLINE OF THE LOMA PRIETA EARTHQUAKE

Since the San Andreas fault and several

active faults run through the California

state, California is one of earthquake prone

regions similar to Japan. The Loma Frieta

Earthquake Is a medium size earthquake with

magnitude of 7.1 along the San Andreas fault. Any particular forecasts for the occurrence

of the earthquake were not made. uses ( United States Geological Survey ) reported that the epicenter was located at

37.036' N. latitude, 121. 883" W. longitude. The

focal depth was ISkin, and this is said rather

deep than normal ones so far observed along

the San Andreas fault. Geodetic surveys

following the earthquake indicate that a slip

at the hypocenter region had a right-lateral

horizontal component eind vertical component

that caused a total oblique-slip displacement

of 2.3 m. The area of the fault surface was

estimated to be 13 km wide and 40 km long.

Since the epicenter was relatively deep, any

fault has not been observed on the ground

surface.

In the epicentral region, due to an extremely strong ground motion ( horizontal

acceleration over 0.7 g ) extensive ground

fissures and slope failures as well as deformations of ground surfaces were developed.

4. STRONG MOTION RECORDS

Intensified strong motion observation

have been made in the State of California by

mostly USGS and CDMG (California Department

of Mines and Geology). During the earthquake,

strong motion records were obtained at 135 stations ( 38 stations by USGS, 93 stations by

CDMG and 4 stations by the University of

California at Santa Cruz).

Fig. 2 shows acceleration records and

response acceleration spectra obtained at city of Corralltos closest to the epicenter ( epicentral distance Is 7km, and the distance

to the fault Is only 1 km). Peak accelerations

are 0.5 g and 0.64 g in two orthogonal

horizontal directions and 0.47g in vertical

direction. The records can be characterized

as the large peak acceleration and relatively

long predominant period of 0.7 seconds. This

may be one of the reasons to cause heavy

damages In San Francisco Bay area.

Fig. 3 shows the distribution of peak

accelerations on ground surface level and

first floor level of buildings. Stations (D to ® located north-east of the city of San

Francisco are on the rocicy bed, and peak

accelerations recorded are in the range of

0.06 g to 0.09 g. On the other hand, peak

accelerations recorded at stations ® to (S) are within the range of 0.16 g to 0.26 g ( 0.33 g at San Francisco International Airport)

which are 2.5 to 3 times larger than those on the rocky bed. It should be noted that

accelerations were largely amplified by the

effects of soft ground.

Fig. 4 shows the relationship between

epicentral distances and peak accelerations

obtained through the analyses of strong

motion data ( excluding the data recorded at buildings and building basements ). in the

figure, empirical formula ( for ground

condition II ) proposed by PWRI'*' is also

shown.

am.x = 232.5X 10°- =3'^x ( A 30)"' (1)

where, am.x. M and A represent peak

acceleration in gal ( cm/sec^ ). magnitude of

earthquake ( 7.1 ) and epicentral distance in km. Since Eq.(l) shows a mean attenuation

derived from a statistical ainalysls of strong

motion records on the ground surface with

moderate to hard soil conditions ( ground

group II specified in the Seismic Design

Specifications of Highway Bridges ). actual

data disperse along the mean value expressed

by Eq.(l). Two curves showing the variation of

a standard deviation (la) from the mean

value are also shown in Fig. 4. This figure indicates that peak

accelerations recorded varies widely even

for the same epicentral disteince. However,

since peal acceleration recorded during the

Loma Prieta Earthquake are within the range

of the mean value ±la , the attenuation with

distance can roughly be explained by Eq.d).

It is often suspected that a special ground

vibration called as echo phenomena occurred

in the bay area because heavy damage was developed not In the epicentral region but in the bay area far from the epicenter. But. excluding the fact that rather long period

ground motion was developed from the fault.

It should be emphasized that the propagation

of the ground motion was not extraordinary

from the viewpoint of the fact that peak

accelerations attenuate with distance in a quite natural way predicted from the past

Japanese experiences.

5. LIQUEFACTION, GROUND FISSURES

AND SLOPE FAILURES

Liquefaction, ground fissures and slope

failures were found at several places.

Damages caused by liquefaction concentrated

in the cities of Santa Cruz and WatsonvlUe

located south of the epicenter, at alluvial

soli ground along rivers and coasts . and at reclaimed lands In San Francisco Bay area.

364
Although various structures were heavily

affected by liquefaction In some areas ( Marina district In San Francisco, Oakland

Port. Oakland Airport, cities of Santa Cruz

and WatsonvUle ), liquefaction did not affect

strongly structures in other districts.

Ground fissures occurred in the

epicentral region. Particularly in the region

about 8kin long and 2.5km wide which is located south-west of the San Andreas fault

and west of State Highway 17, several ground

fissures occurred and caused heavy damages

to some residential houses.

Landslides were also found In several

places in the epicentral region. Some roads were closed to public traffic. Photo. 1 shows

the largest landslide on State Highway 17. It took about one month to completely reopen

this potion to the public traffic.

6. DAi^lAGES TO ROAD FACILITIES

6.1 OLTLINE OF DAMAGES TO ROAD FACILITIES

CALTR.ANS ( Department of Transportation,

State of California ) reported that damage of

road facilities is estimated as $ 18 billion

which corresponds to 22.5% of the total

property damage of S 80 billion caused by the

Loma Prleta Earthquake. The damage to road

facilities includes damages to bridges, cracks

of pavement, falls of rocks and soils, slope

failures along the roads. The damages of

bridges were most serious among them. According to CALTRANS, there are about

1,500 bridges managed by the State and the

almost same number of bridges managed by

counties and cities in the affected region.

Three bridges were critically damaged and

fell down, 9 bridges were heavily damaged. 13 bridges were moderately damaged and 65 bridges were slightly damaged. Damages to three fallen bridges are described in details

in the following.

6.2 Cypress Viaduct (Interstate 880)

The Cypress viaduct consists of a bl-level

concrete decks with length of about 2km on

the Cypress Street of 1-880 through the city

of Oakland. It was designed so as to resist a lateral earthquake force of seismic

coefficient of 0.06 in 1953 and 1954. and its construction was completed in 1957. The

Cypress structure consists of three-span

continuous RC cellular box girders with span

length of 24m. The ends of the girders were interconnected using expansion Joints and

the structure was equipped with cable

restralners at each transverse expansion

joint between the longitudinal box girders.

Bridge columns were made of reinforced

concrete and hinged connections were provided at the bottom of the columns. Three

types of column configurations were employed

as shown in Fig. 5. Type-1 was generally used,

and Types 2 and 3 were employed at the

sections which had a wider lateral beam, such

as ramp and skew sections, and sections to be widened in the future. On Types 2 and 3, lateral beams of upper girder were made of

prestressed concrete using unbonded PC steel

bars.Bridge columns were supported by rafts and pile foundations, and columns and rafts were Interconnected using hinged Joints. The

Cypress Viaduct passed through the

transition area of ground conditions between

reclaimed soft ground, which was a hlnterland-damp ground before 1956, and

well-compacted sand layer of diluvial soli. Soft sUty and clayey layers were piled up almost to 20m In maxlnjum in the reclaimed

zone. The 2-km section of Cypress Viaduct

heavily damaged was demolished soon after

the earthquake. The 1.2-km section of the

Cypress Viaduct between the Interchange 1-80 and the crossing point with 18th street fell down. This section was found to be in the

reclaimed area as shown in Fig. 6. Peak

acceleration in that area was estimated to be

about 0.26 g and the damage should be caused

by the fact that the seismic design coefficient of 0.06 was smaller than the

acceleration actually acted.

Photos. 2-4 show typical damages In which

the upper deck completely fell down on the

lower deck. Fig. 7 shows an estimated

collapsed mechanism of Type-1 column. This

figure indicates that shear failure occurred

by excessive lateral load at lower portion of

hinged Joints which supported upper columns

and girders. At the same time, flexural cracks were developed at the corners between

lateral beams and columns of upper girders,

and then the corners were tore up. Finally

upper decks fell down.

The hinged portion which supported

upper columns was installed at the higher

portion than lateral beam of lower deck, as shown In Fig. 8(a). However, the reinforcement

in the portion was not enough when compared

with that in lateral beam, because small

seismic design coefficient of 0.06 was considered. Therefore, when lateral force was Imposed at the hinge, shear cracks easily

developed. Then when cracks reached to the

axial reinforcement at the ends of cross sec- tions of columns of the lower decks, it was easy to shift to failures as tearing up the

axial reinforcement. Furthermore, fixing

length of U-shaped anchorage reinforcement,

which was arranged to restrict longitudinal

reinforcement at the top portion of columns

for preventing this type of failure, was not long enough. It was also estimated that the

sufficient performance of the portion could

not be expected because of this Inadequate

arrangement of reinforcement at the hinge

portion. On the other hand, the most of

reinforcement at the bottom of upper lateral

365
beams terminated and the reinforcement which was anchored to the columns was almost one

fourth of those in the bottom. The haunch was not provided at the corner of beam-column

connection as shown In Fig. 8(b). Besides, the

reinforcement In the bottom of lateral beam

was not bent down Into the column and

anchorage length was not long enough,

either. Hence, when the bending moment was Imposed at the corner and then flexural

crack reached to the reinforcement at the

bottom of lateral beam, failure occurred at the corner and the reinforcement was easily

pulled out of the column. Therefore, it was

estimated that both of upper columns opened

towards the outside because the two legs were forced to open, and that finally upper

decks fell down onto the lower decks.

6.3 Bay Bridge (Interstate 80) The Bay Bridge is a double-decked

highway bridge with total length of 13.4 km

connecting cities of San Francisco and

Oakland, and consists of twin suspension

bridges between Yerba Buena Island and San

Francisco ( west side ) and truss bridges

Including cantilever structures between

Yerba Buena and Oakland ( east side ). The

bridge was designed in consideration of a seismic design coefficient of 0.1 and Its construction was completed In 1936. A deck span with the length of 15 m fell down on E9 steel pier as shown in Photos.

5

and 6, provably due to out-of-phase

longitudinal vibration between the bridges of

east side ( Oakland side ) and west side ( San

Francisco side ). Although ground soils near Pier E9

consist of mainly silty and clayey layers

which partially Include sand layers, the

surface layer is not particularly thick at the

Pier E9 site. Bed rock can not be found at

least up to 35m under the bottom of the sea. The Pier E9 is of a four-column diagonally

braced steel tower which weis designed as laterally fixed supports for both of a truss

with 154 m long In the east bay side «ind a truss with 88 m long In the west bay side. The

fallen girders for supporting the upper eind lower decks connect two truss bridges on the

Pier E9. One more girder was provided under

lower deck for installing an electric

transformer. The fallen girders were supported by I-beams with a seat length of

12.7cm, and the east side of girders was bolted tc these seats. Although the length

between the end of the girder and the edge of

I-beam was not long enough, fallen girders were designed so that they connect two truss bridges which were fixed at Pier E9. As the

relative displacement between two truss bridges was designed not to be developed, it seems that the falling of the girders during

the earthquake was not predictable.

During the earthquake all of 24 bolts (

</> 25.4mm ) at the Pier E9 connecting shoes of

the east side truss bridge were sheared off

as shown in Photo. 7, and the girders moved

about 14 cm to the east and about 2 cm to the

north. Because of this excessive displacement

the upper deck fell down and struck the

lower deck. Although the shoe seats for the

lower deck were deunaged by the shock of the

fallen upper deck, the girder supporting the

electric transformer prevented the two-layer

girders from falling down completely into the

sea. After the earthquake, two damaged decks were removed and new girders were designed

and erected. The residual morements of 13- span truss bridges ( total length of 1,140 m ) were replaced, one by one, using Jacks at each span. Although the restoring work of

the girders was estimated to teike at least a few months, the work was completed and

reopened to public trtifflc on November 17, Just one month after the earthquake. A great

deal of efforts of CALTRANS should be highly

applauded.

6.4 Struve Slough Bridges (State Highway 1) Struve Slough Bridge, crossing a damp

soft soil zone located on State Highway No. 1 south of WatsonviUe, Is a 21-span continuous

RC girder bridge. Three gerber hinges ( expansion Joints ) were located In each

bridge. Although the abutment at both ends

of the bridges were supported on stiff

ground, piers were built In the damp soft soil

( ground water level of -0.8m ). The piers

consisted of pile bents ( # 38 cm ) which were

fixed to the lateral beam of the deck. The

bridges were designed in 1961 according to the former seismic design code, and the

construction was completed in about 1965. As shown In Photos. 9 and 10, the center

section about 100 m long of the bridge, which

has a total length of about 232 m, moved

residually about 80 cm in maximum In the

transverse direction and fell down. Four

piles punched out through the deck slab and

the head of piles stick out from the slab as long as 1.2 m to 1.8 m. Photo. 11 shows the damage to the

connection of piles to lateral beam. The

heads of the pile bents completely failed and

the some axial reinforcements (6x0 20mm ) were seriously bent and broken. The pile

bents were completely failed at their heads

and dislodged from the lateral beam, crashed

to the deck slab, and seriously Inclined.

Some pile bents completely fell down onto the

ground. The 4 pile fcents punched out through

the deck slab by the falllng-off energy of

the deck slab as shown in Photo. 10, which

Implies that vertical bearing capacity of the

piles was estimated high enough. Since a gap

about 60 cm wide in the transverse direction

between piles and ground surface can be

found. It seems that the piles largely moved.

366
Although the portion of the pile over the

ground level was made of reinforced

concrete, portions under the ground level are estimated to be cast-ln-place plain

concrete piles, so called dropped-ln-shell

piles which are one type of Raymond piles.

This type of pile has not been employed in Japanese bridges.

It is considered that an excessive

earthquake lateral force over the one

considered in the design may be a major

reason for such a critical damage. It is also

considered that the ground acceleration was

at least more than 0.2 g at the site, and that

the substructure consisted of 4 pile bents,

which connected to lateral beam, could not

resist the lateral force in the transverse

direction. Furthermore, it is supposed that

the plain concrete portions of the piles

under the ground level failed during the

earthquake. Judging from appeared damage of

the plies.

7. DAMAGES TO LIFELINE FACILITIES ( EXCEPT

ROAD FACILITIES )

This section briefly summarizes damages

to lifeline facilities ( except road facilities

above mentioned ) such as water supply and

sewer facilities, electric communication

facilities, electric power supply facilities, gas supply facilities, broadcasting facilities,

data communication and computer facilities,

harbor and airport facilities, subway, and

river facilities. The serious damages to water supply and

sewer facilities were not developed except some damages at Marina District and South

Market area in San Francisco.

Although slight damages to cables,

antennas of automobile phones and electric

power apparatus were found, physical damages

to electric communication facilities, were not

found, either. However, since a large number

of phone calls from not only domestic but

also world-wide countries rushed to the

region { 7 times for receiving and 2 times for

sending, when compared with the ordinary

numbers ). congestion of the phone

comjDunlcation continued for two days because

of exceeding the dealing capacity by far. Communication companies In the affected

region restricted a considerable number of

calls ( especially for receiving ) for two

days.

Electric power supply stopped for a day

or two mainly in some areas In San Francisco,

due to damages to power generation,

transformer sub-station, and power

distribution facilities. The power failure

seriously affected the normal living of the

residents, and interrupted functions of various facilities, which depend on electricity. Gas supply facilities in three cities close

to the epicenter and some other areas ( with

soft reclaimed ground ) In San Francisco were

damaged. In particular, cast-ironed pipes were damaged due to liquefaction of ground

in Marina District in San Francisco. The gas mains of total length of about 16 km and

distribution pipes to houses were renewed

after the earthquake.

Although broadcasting facilities were

affected by the power failure, direct damages

were slight and therefore broadcasting

continued. It seems that the continuation of

broadcasting greatly contributed to security and ease of the residents.

Electric power failure and damage to buildings caused losses of functions of data

communication and computer systems In San

Francisco. However, little damages to physical facilities were found.

Subsidences of grounds and damages to buildings of harbor and airport facilities,

mostly caused by liquefaction of ground, were found In Oakland Harbor and San Francisco

International Airport.

Although the BART subway ( including a section of submerged tunnel ) operated in San

Francisco Bay area were affected by the

power failure, and a little increase in leakage of water at approach sections of the

tunnel was observed. Operation of the subway

restarted within a few hours after performing

quick Inspections and temporary repair

works. The BART subway as well as ferryboats

played an Important role for public transportation until Bay Bridge reopened to public

traffic. Several damages to river embankments,

such as subsidences, were found in the

eplcentral region. Minor cracks at the sur- faces of upper and lower sides of earth dams were also found.

8. MEASURES AGAINST DISASTER AND VOLUNTARY

ACTIVITIES

Twenty two fires occurred in San

Francisco. No fires were big. In the worst

case 4 houses were burned down. It should be

recognized that usual disaster drill and

education on earthquake disaster for the

residents prevented big fires and occurrences of panic. In despite of the power

failure, the public order was maintained after

the earthquake, with helps by volunteers.

Therefore, no critical problems happened to public security or traffic flow. Emergency measures against disaster were

Implemented by the federal, state and local

governments In the sajne buildings ( In some

cases ). Each role of these organizations was shared clearly. Various measures for the sake

of suffered people were conducted.

it should be particularly emphasized that

people of the U.S. have voluntary spirits

since the foundation of the nation, and that

the voluntary activities by private sectors

367
such as the Red Cross and the Salvation Army

played Important roles in reducing disasters.

This Is an Important suggestion to Improve

Japanese situations of measures and policies

against disasters.

9. CONCLUDING REMARKS

Although the Loma Prleta Earthqual<e was medium In size with magnitude of 7.1. San

Frainclsco Bay area was heavily affected. This

happened since the great San Francisco

earthquake of 1906 ( magnitude of 8.3 ) which

affected most seriously in the history of

U.S.A. It Is said that the Loma Prieta Earthquake w£is a first experience in which a medium size earthquake struck a modern

highly information-oriented and densely populated city.

The earthquake indicates how seismic

performance of lifeline facilities ( such as road. electric power and Information

communication facilities ) is important. The

earthquake caused fatalities of 62 lives and

heavy structural damages. However, the ef- fects of the earthquake have not ended yet

and will happen in various forms. In California state, sales tax has been raised as much as 0.25% for 13 months of the period

from Dec. 1, 1989 to Dec. 31. 1990, and an increase In state revenue of $8 millions is expected to spend for restoring.

Furthermore, federal government will spend

the ordinary budget of 3 times as much as state one for the purpose of restoring public

facilities ( naunely. federal and state governments shared 75% and 25% of the budget

for the restoring public facilities,

respectively ). It is indicated that frequencies of occurrences of large earthquaikes In Japan

are higher than those In U.S. Although it is not appropriate to directly apply the

experiences of the Loma Prieta Earthquake to Japan ( because seismic design methods of

structures including the seismic forces are not same ). it Is essential to consteintly promote and Improve measures against disasters

caused by large earthquakes.

10. ACKNOWLEDGMENTS

Great supports and efforts were given,

from both U.S. and Japan, to the Japanese

Government Study Team for the Loma Prieta

Earthquake. The author, on behalf of the team members expresses our appreciations to following organizations and Individuals

From the Japan side, the teaim got a great

deal of supports and cooperations from the

Ministry of Foreign Affairs,

Consulate-General of Japjin in San Francisco.

Disaster Prevention Bureau of National Land

Agency ( especially Mr. Ichiro Ichlkawa, Chief

of the Bureau ), Ministry of Construction,

Japan Highway Public Corporation.

Metropolit£Ln Expressway Public Corporation

and Honshu-Shlkoku Bridge Authority. The

author would like to greatly appreciate to the concerned persons of the above

orgemlzatlons. And Mr.Hldeakl Oda. Sub-leader

of the team, and other members of the team

are acknowledged for a great contribution to field investigations and preparation of the

report. A great deal of appreciations are also

directed to Dr. R.N.Wright ( Dlrecter of

Center of Building Technology. National

Institute of Standards and Technology and

also U.S. Chairman of the Panel on Wind and

Seismic Effects. UJNR). Mr.N. J.Raufaste ( Sec- retary-General of the Panel. NIST ). Dr. H.S.Lew ( member of the Panel. NIST ), Dr. A. G.Brady ( member of the Panel. United

States Geological Survey, Department of the

Interior ), Mr. J. D.Cooper ( member of the

Panel, Federal Highway Administration ). and

Mr.J. H.Gates ( member of the Panel. CALTRANS)

for providing a great deal of supports with

regards to 1) Invitation of the team. 2) arrangement for the field investigation.

3)Jolnt investigation of the affected structures, and 4) offer of many information

Including valuable photographs and slides.

11. References

1) Japanese Government Study Team for the

Loam Prieta Earthquake ( T. Iwasakl, H. Oda.

et. al. ) : Report on the Damage by the Loma

Prieta Earthquake, Disaster Prevention

Bureau, National Land Agency, January, 1990 ( in Japanese ). 2) Iwasakl, T., Kawashlma, K.. Itoh. Y., Aklmoto, Y. «ind Tatsumi, M. : Reconnaissance

Report on Damage by the Loma Prieta

Earthquake, Civil Engineering Journal.

Vol.32. No. 2. February. 1990 ( in Japanese ). 3) Iwasakl. T. find Kawashlma, K. . Damages to Bridges during the Loma Prieta Earthquake,

Journal of Roads, No. 1990-2, February, 1990. Japan Road Association. ( in Japanese ) 4) Kawashlma, K.. Aizawa. K. and Takahashl. K.

: Attenuation of Peak Ground Acceleration,

Velocity and Displacement Based on Multiple

Regression Analysis of Japanese Strong

Motion Records, Earthquake Engineering and

Structural Dynamics, Vol.14, 1986

5) Lew, H.S. (editor). Performance of

Structures during the Loma Prieta Earthquake

of October 17. 1989. NIST Special Publication

778. NIST. January, 1990.

368
Foil of Decks of

I Boy Bngge. 1-80

Richmond-Rafoei Bridge

CracK'^ of Bridge °iers of EmbarcaderoViaauct, State Freeway 4S0

i Damages to Lifelines and Houses

1 of Morino Disf net

Golden Gate Bridge

CracKs of Bnage Piers of US

National Hign way No i

rocks of Bridge Piers >280^^^-^-

Viaduct neor Mission Creek :—DaiyCitj

I Collapse of Cypress

' Viaduct, 1-880

Cracksof Bridge Piers 1-280

Vioaucf near Selby Street

Pacific Ocean

Legend

" -ia'S Interstate Frwy C" U 5 Notional Frwy

5 State Frwy

DfinciDOl Hwy

Crocks at Lexington Earth Dam

Slope Foil ure along

Stote Frwy No 17 Aftershocks

Settlement of Dikes of Son Lorenzo River

Collapse of Struve Slough

Bridge on Stote Frwy No

Settlement of D ikes and CracKs of Bridge Piers

ot Pojoro River

Monterey County

30krr

Fig. 1 Map of Affected Area

369
600

300

0 -300

-600

600 r 300-

0

-300

-600-

600

300

0

-300

-600

HorizontQi-A ( Qmo. = 469 gal

,

h^-*

Vertical (Omo« = 43 I gol

)

Horizontal - B (Omoi = 6l7gal)

(A)

Fig

5 10 15 Time (sec)

Acceleration Wove Forms

20

- 5.000 ^ 3.000

2.000

1 1.000 ^ 700

= 500

i 300

I 200 ^ 100

Horizontal - B (amo.= 617 gol ) - - "s. Horizontal -A

____><\(amo. = 469gal)

- Vert ical

• (a moi= 431 gal )

: Damping Ratio \ \A ;

h = 005

50

30

0.05 0.1 02 03 05 07 1 2 3

Noturol Period (sec)

(B) Response Acceleration Spectra

Acceleration Wave Froms and Response Acceleration Spectra

(for Damping Ratio of 5%) Recorded at Corralitos Closest to the Epicenter (from CDMG)

Fig, 3 Distribution of Peak Accelerations on Ground Surfaces and

First Floors of Buildings

(from uses and CDMG)

07

0.5

03

02

0.07

0 05-

0.03-

0.02

0.01

\

-° •o \o

Legend J) \ \

Peak Acceleration Measured \ o On Alluviol Soils \^ \

• On Rocks °

Empiricol Formulo Proposed by PWRI (for Soils) — Mean Values (m)

Mean (m) i Standord Deviation ( CT)

Fig

5 7 10 20 30 50 70 100 200 300

Epicentral Distonce ( km )

A Relationship between Epicentral Distances and Peak Accelerations

370
(•2.0119

J H mae Hinge

|^;0,10-0.14g

iHinge

Y'X' 0 10- 0 14 g

Hinge

(a ) Type

hinge

Hinge 0Ol7g

P^O.Zlg .2.^0.2 lg- JHinge

Hinge

Legend

Reclaimed Land ! Formerly Bay )

Recloimed Land (

Mentt Sand :

Formerly Marsh

)

Sedimentory [ ^ ueuo uepositj

- Coastal Line as

luviai Loyer

of 1856

• B-19- 34 Boring Numbers

1- 80 Domolished

\ Section

(b) Type 2

Hinge

Tl

Hinge 0 I9g

3t

^O.I7g

Hinge

Hinge

^^0.29g

Hinge

Son Franc

(c ) Type 3

Three Typical Types of Piers

of Cypress Viaduct (Accelerations;

indicated correspond to those

which cause failures at respective

locations, estimated by Prof,

Priestley, Univ. of Calif, at Fig. 6 Location and Geological

San Diego) Conditions of Cypress Viaduct

. Upper Girder

Lower Girder

Cypress Street

Footing —

J |.

,. Pile -^11

(Al Crocking ot Hinges

^ Upper Girder

4 Foil

-rock

Spoiling of concrete ond

exposing of reinforcement

Detail of Hinge

(B) Collapse of Hmges !C) Fall of Upper Girder Fig. 7 Estimated Collapse Mechanism of Cypress Viaduct

(Type 1 Pier)

371
Upper Column

Lower

Lateral Beam

14D32mm Lower 'Column

( Longitudinoi Bars)

Collapsed Section Dl5mm

Confining Bors) Upper

Loteral Beam

44D 57mm

Longitudinal

Bors)

4 D 32 mm

/Hinge N Connection

\ Bors ^

4Dl3mm . 0.9m

(Hoop)

< . . .N 1 E

30 D57mm

Longitudinal

Bors)

8 m

. Dl3mm

Hoop

(Pitch = 30cm)

(Nil

2 m

SectionA-A Section B-B Section C-C

D 16 mm

(Confining Bars)

5 D 57 mm

~1 ( Reinforcements

C below Beam )

D 1 3 mm

Hoop ( Pitch = 30 cm

A) Hinge at the Bottom of Upper

Column, Lower Laterol Beom , ond Lower Column

(Bl Upper Loteral Beam ond Beam-Column Connection

Fig. 8 Structural Details of Hinge anci Beam-Column Connection

at Cypress Viaduct

To Son Froncisco To Ooklond

Upper Rood Surface . -- Upper Dec k Details of A

__Fixel Third Deck for tronsformerp

IS located at this level

[After EorthquoKe

Upper Deck

Fixed

j

127sm"

Seat Length

24 Bolts, 0 25 4 mm

All bolts were sheared Off. ond the girder moved obout 14cm toword Oakland Side

Detoils of B

After Reooir

!

20 3 cm

Seat Length

Fig. 9 Fall of One-Span Decks on Pier E9 of

Bay Bridge and Outline of Its Repair

Effective Width 8.5m

.21

Section A-A

RC Piles obove Ground Deformed Axiol Bor, * 20mm

Round Bor Hoop , 4 mm

1

I UI'TTT —I

J T-r-| JI—'

'-'

(B) Failure of Pile Bents

Fig. 10 Damage to a Pier of

Struve Slough Bridge

372
Phote. 1 Slope Failure along State Freeway No. 17 (One of Largest Landslides caused by the Earthquake

with Height == 20m and Length = 100m; It took one month to completely open this portion to the public

traffic. This photo was taken in the end of November,

1989 after repair works finished.)

Photo. 2 Collapse of Cypress Viaduct on 32nd Street, Oakland. Typical damage to Type 1 piers (from

CALTRANS)

.

Photo. 4 Collapse of Type 2 pier of Cypress Viaduct.

Upper columns fell down due to failures at upper

hinges. Reinforcements at the bottom of columns are

seen. Unbond PC bars are seen atop (from CALTRANS).

Photo. 5 Fall of one-span decks on steel pier E9

.

Upper and lower decks fell down, but fortunately those

two decks stopped on the third deck which install transformers (from CALTRANS).

Iso.vgfEle.c-k' umnB^mfwTumn

Photo. 3 Collapse of Type 1 pier of Cypress Viaduct.

Upper Deck completely fell down onto lower deck due

to failures at hinge at the bottom of the upper column

and at beam-column connection (from CALTRANS).

Photo. 6 Removal of fallen decks. Bearing supports

for upper deck are seen. The length of bearing

seat is 12.7 cm (from CALTRANS).

373
Photo. 7 Movement of the fixed bearing support on the east (Oakland) side of pier E9. Bolts were shear

off, and the support moved about 14 cm to Oakland

side. Due no this movement the upper deck (with

bearing seat length of 12.7cm) dislodged from the supports, and fell dcuirn (from C.\LTI<^\NS).

Photo. 8 Due to the impact of the fall of the upper deck onto the lower deck, the lower deck fell down

onto the third deck which install transformers

(from CALTRANS)

.

4 Piles Punched out through Slab

Photo. 9 Overall view of Struve Slough br iOKc, RC girder bridge with 21 spans and about 232m long,

crossing over Struve Slough. There are 3 girder

hinges along the longitudinal direction.

Photo. 10 4 Pile bents punched out (up to 1.2 to 1.8m)

through slab at Struve Slough bridge.

Photo. 11 Heads of pile bents completely failed and dislodged from the lateral beam. The superstructure

moved about 80cm in the transverse direction (left

in the photo . )

.

Photo. 12 Pile bent largely moved, causing a big hole surrounding the bent. Peat soils of high water content with high water table of 80 cm below the ground surface look very soft.

374
Manuscripts Authored

for Panel Meeting but

Not Presented Orally

Experimental Study on Lateral Flow of Ground

Induced by Soil Liquefaction

by

Yasushl SASAKI^ Kcn-lchI TORIDA' nideo IHATSUMOTO'" and Sholchl SAYA > « Z 1

SUMMARY

A series of shaking table tests are carried

out on the lateral flow of the ground caused

by liquefaction. The nodel grounds are 6.0

long, 0.8 D wide and their average height is 1.2 . The slopes of ground surface and the lower boundary condition of liqueflable layer

are varied. The deforsation of nodel grounds,

excessive pore water pressure of the liquefied

layer, and the acceleration of the grounds are monitored. Based on the test results, the

Influence of the slope, the input lotion and

the thickness of liquefied layer to the ground

flow is examined.

KEY WORDS : Liquefaction, Sloped Ground,

Ground Flow, Shaking Table Test

1. FORWARJ)

The soil liquefaction has caused severe

damages for structures during the past

earthquakes in Japan. Haaada et. al. (1986)

pointed out that the fairly big aaount of lateral flow of ground could be seen broadly

in the areas in Niigata and Koshlro where the soil liquefaction of the ground apparently

took place during the Niigata Earthquake and

the Nihonkai-chubu Earthquake respectively.

Therefore it becomes necessary to consider two types of effect induced by soil liquefaction

for the earthquake resistance design of structures; the loss of bearing capacity of ground and the Influence of ground flow. However the nechanlsm of ground flow and

predicting method of this phenomena are not yet clarified at present.

Under these circusstances , the authors conduct

a series of shaking table tests to study the mechanism of ground flow and to establish the procedure for estimating ground flow. This

paper describes the essential factors causing

lateral flow of ground due to liquefaction

from the experiments.

2. MEXnOD OF EXPERIMEKTS

Shaking table tests on eight model grounds

(Model-1 — 8) are carried out. The grounds

except for Model-7 and Model-8 basically

consist of three layers. The lowermost sand

layer is compacted enough not to liquefy

during shaking, the middle layer is loose sand which is expected to liquefy, and at the top

of them is set the unsaturated sloped

uppermost layer. The Model-7 dose not have the uppermost layer. Instead the lower boundary of liqueflable layer is Inclined. The ModeI-8

does not have the lowermost layer. Eight

models are shown in Fig. 1 and Table 1. The thickness of liqueflable layer the slope of surface ( 6 ,) and lower boundary of liqueflable layer ( d o ) are varied to have

different conditions in order to examine the influence of these factors to the lateral

flow. Furthermore, for Model-5, 6 and 8, shot

of lead or gravel is used as the unsaturated

uppermost layer to give the larger overburden

pressure and lesser cohesion. The water level

is set equally to the upper boundary of liqueflable layer. The characteristics of the materials for ground models are indicated in Table 2. In each test, constant sinusoidal acceleration

with a frequency of 2 Hz is employed for 20 seconds (40cycles). and several levels of maximum acceleration are applied stepwise to the shaking table as an input motion. The

measured maximum accelerations of the shaking

table and the results of lateral displacement

at surface are summarized in Table 3. As shown in Fig. 1, accelerometers , pore water pressure meters, strain gage meters and

displacement meters are installed in or on the ground models to measure the time history of acceleration, pore water pressure and displacement of layers during shaking.

Vertical marked lines are also installed at the side of each model grounds in order to monitor the deformation of the layers.

• Directer. Earthquake Disaster Prevention

Department, Public Works Research

Institute, Ministry of Construction

Head, Ground Vibration Division,

Earthquake Disaster Prevention

Department, ditto

••• Research Engineer, Ground Vibration

Division, ditto

Ground Vibration Division, ditto

377
3. RESU1.TS OF EXPERIMENTS 3.3 RELATIOH BETWEEN DURING AND AFTER

From the above described shaking table tests,

principal characteristics of lateral ground

flow due to liquefaction and factors

concerning to the ground flow are drawn as follows

.

3.1 DISTRIBUTION OF GROUND FLOW

Fig. 2 shows the typical distribution of lateral displacenent at the surface (Model-4

and 6). From this figure, it can be seen

that tensile strains occur at the upper part

of slope (right hand side in Fig. 2). the lower part of slope seeas to be coipressed.

Fig. 3 shows successive change of vertical

marked lines during excitation at the Step-2

in the cases of Model-4 and 6. This figure is drawn from photographs taken through the transparent glass window on the side of the container. Based on this figure, it can be

noted that the deformations of ground surface

is Induced by the deformation of the liquefied

layer but not by the deformation of the uppermost layer. The displacement at the lower boundary of liquefied layer is almost

zero, and increases upwards in the liquefied

layer

.

3.2 RELATION BETWEEN ACCELERATION. EXCESSIVE

PORE WATER PRESSURE AND GROUND FLOW ON TIME

HISTORY DURING EXCITATION

Fig. 4 shows the typical relation of time

histories between acceleration, excessive pore water pressure in the liquefied layer and

lateral displacement at the surface. at the

Step-2 In the case of Model-6. From this

result, it can be seen that the acceleration

in the llqueflable layer becomes spiky Just before the excessive pore water pressure in llqueflable layer approaches to the effective

overburden pressure, and decreases after the

complete liquefaction in the layer is obtained

.

On the other hand. Fig. 5 shows the relation of time history between the lateral displacement

at the surface and the excessive pore water

pressure in liquefied layer at Section A. B. C and D respectively. From this figure, the displacement at the ground surface begins to Increase gradually Just after the excessive

pore water pressure ratio ( ^r) u/ a 'v) approaches to the range of 0.6 to 0.8. and

also continues to Increase gradually during

excitation after the complete liquefaction in the subground layer is attained.

EXCITATION

Fig. 4 also shows the behavior after stopping

excitation. After excitation, the excessive

pore water pressure keeps the pressure

by seepage pressure for about 2 minutes, and

It takes about 20 minutes for high excessive

pore water pressure to decrease to almost

zero. On the other hand, the displacement at the ground surface Increases a little after

excitation, but it Is very little comparing

with that during excitation. From the site

observation on the liquefaction and lateral

ground flow in the past earthquakes, sand

bolls, water springs and ground movements

are reported to continue for a while after

the earthquake. The cause of the difference

on ground flow induced by liquefaction

between the actual phenomena at sites during

earthquakes and experiments can not be cleared

at present.

3.4 RELATION BETWEEN INPUT MOTION AND GROUND

FLOW DIRECTIONS

Fig. 6 shows the direction and quantity of ground flow at the surface measured by the bench mark points, against the direction

of input motion, at the Step-2 in the case of Model-8. From this flg^jre. the surface of the semi-corn shaped slope and the neighboring

horizontal ground seems to move almost

uniformly and radially, that is, in the direction of slope.

Judging from these characteristics concerning

the ground flow, it seems that whichever the direction of the excitation is. the liquefaction induces the ground flow towards

the direction of the initial shear stress

in the llqueflable layer enforced by sloped

overburden load.

3.5 EFFECTS OF SURFACE SLOPE AND LOWER

BOUNDARY SLOPE

Fig. 7 shows the influence to ground flow at the surface by slopes of both surface and

lower boundary of llqueflable layer. As seen

in the figure, the influence of the slope of the lower boundary of llqueflable layer to the ground flow seems very small. However, when

both ground surface and lower boundary of llqueflable layer are sloped as the case of Model-3. the slope of the lower boundary of llqueflable layer is likely to increase the ground flow.

378
3.6 ESTIPUTIOW OF GROUyP FLOW

FroB the above mentioned test results, It

is reduced that the slope of surface { 6 ,) , the thickness of liquefied layer where the

excessive pore water pressure ratio ( z3 u/

(7 'v) is almost equal to 1.0 (H) and the

duration time of excitation while the ratio ( ^ u/ 'v) is almost 1.0 (T). can be

considered as the main factors relating to the

quantity of ground flow Measured at the surface (D). Fig. 8 shows the relation

between the parameter D/(T H) and the slope

of surface( e ,) in the Sections A, B. C and D at the Step 1 and Step 2 for all Models

excepting Model-8. Based on this figure, it can be seen that the parameter D/(T H) has a tendency to Increase according to the increase

of the slope 6 ,.

CONCLUSION

Based on the above-mentioned results for shaking table tests, the following can be concluded

;

(1) The pore water pressure Increases firstly

during excitation, preceding the lateral

ground deformation. and the lateral

displacement of the sloped ground begins

to increase remarkably when excessive pore water pressure ratio in the underlying

layer exceeds about the range of 0.6 —

0.8.

(2) The lateral deformation occurs only In the liquefied layers, and the surface

layer is displaced according to the deformation of underlying liquefied

layers

.

(3) The ground seems to be deformed by shear

force during excitation and the deformation mode of ground in the

direction of depth differs according to the position and the lapse of exciting

time

.

(4) Just after stopping the excitation, the excessive pore water pressure In the liquefied layer keeps the pressure by seepage for a while and begins to decrease

gradually. On the other hand, the lateral

ground flow almost stops.

(5) The lateral ground flow is affected

severely by the slope of ground surface,

and however is very small in case that the ground surface is flat even though the lower boundary of the llquefiable layer is sloped.

(6) The direction of excitation has no significant influence to the lateral flow

of ground, and the lateral ground

deforiation is related to the direction of the slope of surface.

(7) The lateral ground flow at the ground

surface in this series of experiments can

be expressed by the slope of surface,

thickness of perfectly liquefied layer and

the duration time of excitation after

almost complete liquefaction.

As mentioned above, characteristics on ground

flow Induced by liquefaction are clarified and principal factors which cause liquefaction

induced ground flow are qualitatively

summarized by Fig 8.

REFERENCES

II Hamada, M., Yasuda, S.. Isoyaraa, R. and

Eioto, K. "Study on Liquefaction Induced Permanent Ground Displacements,"

Association for the Development of Earthquake Prediction, 1986.

2) Sasaki, Y., Matsunoto, H. and Saya, S. "Shaking Table Tests on Ground Flow

Induced by Liquefaction", Technical

Memorandum of Public Works Research

Institute. No. 2768, February, 1989 (in

Japanese)

.

379
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o u

CO

at U u I W 3 U — O- >» js — n

-J

o

o ^ « a s 9

o s 9 _ o —

(>0 CB _J

O CJ -V

ft n

(/) 1/3 — o

u as

in

LA in lA

PJ CO PS

O'O U5 O O O lA

lA . LA lA . ^ O

h> C4 O

-« N eo la CO {>»

lA

00

CO

CO

CO o

c

a; E E

•1—

(

Q-i u

o &

a.

CO K

UJO

CO

u.o

o 0-1 03

(0 CO u 3

.a

lA CO (O lA fx* O) CO r«> CO O) lA N m

N o o ^ o —

X c c n "o U

s O

in o o o a s N s

V » M n o B M aa N V > S O n

u n U e B

ai u M

u

3 3 3

T5S B n H S M V

to rt <« « o

to lA

«n U

o

u

eu

u> o

*» i:: o

SI > B

U ra B u

o u o 3 >

o w B n

V re h> H

r<9

380
Table 3 Input Motion to the Shaking Table and Lateral

Displacement at Surface after Excitation

Model Shaking MaziiuR Acceleration Lateral Displacesent at

No. Step No. at Table a nai (gal) Surface (Section B) ()

1 8 0 6 1

1 2 1 1 nU

11 6 9

3 1 D U 1 1 8

4 U 5 6

1 6 5 2 8

2 2

1 1 1 1 n 4 4

3 1 5 5 1 6 5

4 1 0 5 6 2

1 6 5 6 8

3 2 1 1 0 1 5 6

3 9 0 8 7

4 8 5 3 5

1 6 5 6

4 2 9 5 7 2

3 1 4 0 1 4 2

4 1 9 0 4 4

1 6 0 1 5 2 1 0 1 4

3 1 5 2 2 1 0

4 1 5 4 6 9

1 6 0 1 6 2 1 0 4 1 0 1 3 1 5 3 1 0 6

4 1 5 4 5 4

1 6 4 2

7 2 1 0 3 1 7

3 1 5 4 1 1 4 1 5 1 6

1 5 1 1 4

8 2 9 4 1 1 3 9 5 5

381
^^^^^^^^ ^^^^^^^^^^ ^ + 't + **^ t o' o

||8

t

[ ^^^^ 0 !Bi J

t t a.

21

<» ••». * T "7 sH ia-U«illiU< Mo d • I - 1

to4 ri-

•

.-..:.....<;.- al «

oi ' "4 Mo d e

I -2

1 ^.—^

III lit

. r

SI Mod«l-3

, tt'it — —

it

^!

laf

"oT

Mo d • 1 -4

PL

-fm .

:

. S

,. -lij . V :»^-: • -SJ . «

« • »k>OX fl Tri—' ' ' ai—

21 Mod. 1-5 21

tt-ox , . 1 -t rS|—

1

St .

: ^' V;' st rirL

" •••OX • •

si 2: ^

21 Mo d •

1 -6

' J J M-OI J J

21

r »;—iim

•#^»- '

e4

SI mo looo ,,, 1000 1000 1000 \m • 1 Mo d • I -7

?0«200>i000

MO 400 <00 '00

• Ace«lirM4itr [

•FitMit Mark Filii | ••Siriit &4f> Utttr I III it« liTtri I

_ / Di rtot ion \

Mo d o 1 -8

Fig. 1 Ground Models and Measurements

382
0 1 2 3 4 5 ^6

Tension Lateral Position from the Left Edge of the Container |m]

Compr«ssion (a) Model-4 ( 0 s = S% , e b = 05j, Hl = 35cni)

0 1 2 3 4 5 6 Tension Lateral Position from the Left Edge of the Container [m]

Compression (b) Model-6 ( 0 s = 0 , 55^. . 0 b = 0%, Hl = 35cm)

Fig. 2 Lateral Deformation of Ground Surface Resulted from

Bench Mark Points

383
384
3 ;: :5 :3 := 2C (a) Lateral Displacement of Surface

'J •lOO'- u< - JOC- -iOO-

- 5

= ;c :o :5 v- (b) Pore Water Pressure of in the middle

Liqueflable Layer

V ic :5 :o :5 ; (c) Response Acceleration of Liqueflable Layer

.3 ;s 20 :s 30 (d) Acceleration of Input Excitation

Fig. 4 Relation Between Lateral Displacement of Surface,

Pore Water Pressure and Acceleration of Liqueflable Layer

(Model-6, Step-2, Section-A)

Fig. 5 Relationship between Pore Water Pressure Ratio and Displacement

of Surface (Model-6 , Step-2 : a inax = 104gal

)

385
Direction of Exc i t a 1 1 on

200 100 * HORIZONTAL**-

0 100 200

SLOPE —- * HORIZONTAL *

(a) Direction of Ground Flow

A B C D E F G H I Section in the Direction of Ci rcumt erence

(b) Distribution of Ground Flow

Fig. 6 Relationship between the Direction of Ground Flow and

the Director of Input Motion (Model-8,Step-2: a max=94gal)

386
0 12 3 4 5 ^

Tension Lateral Position from the Left Edge of the Container [a]

r 14 T' r _" ^ u; 1 ^^^^^ L . L . . , _ 1. - ._ J .. ....

.f 10

0

I

-10

Cofipression

Fig. 7 Influence by Slopes of Ground Surface and Boundary of

Liquefiable Layer for Ground Flow

0 2 4 6 8 10 12 14 Slope of Ground Surface 9z \%]

Fig. 8 Relation between Lateral Flow and Slope of Surface

(Section A.B.C.D)

387
Development of Advanced Reinforced Concrete Buildings

Using High-Strength Concrete and Reinforcement

by

Tatsao Murota \ Hisahiio Hiraishi Talcaski Kaminosono , Ma^aomi Teshigawara and Hitoshi Shiohara * ABSTRACT

This paper lists research subjects conducted in 1989 in MOC's five-year research project on reinforced concrete

buildings using high strength and high-quality concrete and reinforcing steel. This project is simply referred as

to "New RC. The paper also describes the outlines of some of important research subjects related to structural performance.

1. INTRODUCTION

Reinforced concrete (hereafter referred a5 to "RC") has

widely been used for medium scale buildings because of the low cost, e.xcellent durability and easy maintenance,

so on. However, it is impossible to realize the super highrise buildings and buildings with long spans which are currently required from the social point of view, if the material strength remains within ordinary ones. The Project aims to produce high strength concrete of the

specified strength of from 300 kgf/cm^ - 1 ,200 kgf/cm^ (30

- 120 MPa) and high strength and high quality reinforcing

steel bars of the yield strength of 4,000 kgf/cm^ - 12,000

kgf/cm- (400 to 1,200 MPa), and to develop new fields of RC buildings by utilizing them.

2. RESEARCH SUBJECTS CONDUCTED IN

1989

The following researches were conducted in 1989.

1. Researches on Concrete

(a) Quality e.xamination of cement for high strength

and super high strength concrete

(b) Quality e.xamination of the aggregate for high

strength and super high strength concrete

(c) Quality standard and application criteria of the admixture for high strength and super high

strength concrete

(d) Development of the binder for high strength

and super high strength concrete

(e) Design method of mLx proportion for high strength

and super high strength concrete

(f) Evaluation method of workability of high strength

and super high strength concrete

(g) Evaluation of mechanical properties of high strength

and super high strength concrete

(h) Evaluation of long-term properties of high strength

and super high strength concrete

(i) Evaluation of compressive strength of concrete

in structures

(j) Evaluation of durability of high strength and

super high strength concrete

(k) Development of essential construction techniques

for high strength and super high strength concrete

(I) Evaluation of fire resistance of high strength

and super high strength concrete (m) Evaluation method of quality reliabQity of high

strength concrete in structures

2. Researches on Reinforcing Steel Bars and Confined Concrete

(a) Production techniques of high strength and su- per high strength reinforcing steel bars and their mechanical properties

(b) Development of standard hooks for lateral re- inforcement

(c) Effect of lateral confinement on compressive

behavior of columns

(d) Behavior of columns subjected to axial force and bending moment

(e) Analytical modeling for stress vs. strain rela- tions of concrete and steel, and bond characteristics

(f) Development of computer program using finite element analysis method

(g) Yield criteria of concrete under biaxial compression

(h) Deterioration characteristics in compressive ca- pacity of cracked concrete

(i) Tension stiffening of cracked reinforced concrete

(j) Anchorage strength of longitudinal reinforcing

steel bars in beams bent inside exterior beam

column joint panel

(k) Bond characteristic of longitudinal reinforcing

steel bars in beams and columns

3. Researches on Structural Performance of Members

(a) Deformation capacity of column under highaxial force

(b) Deformation capacity of Beams

(c) Bond splitting failure of beams beyond flexural yielding

(d) Flexural behavior of shear walls

(e) Shear behavior of shear walls

(f) Shear capacity of beams

'Director, Structural Engineering Department, Building Research Institute 'Head, Structural Division, Structural Engineering Department,

Building Research Institute ^Senior Resarch Associate, Large Scale Structure Testing Division, Production Department, Building Research Institute "Research Associate, Structural Division, Structural Engineering Department, Building Research Institute

388
(g) Shear capacity of columns

(h) Shear capacity of interior beam column joint

panels

(i) Shear capacity of exterior beam column joint

panels

4. Researches on Structural Design

(a) Structural design examples and their discus- sions

• A sixty-story frame building and that with

shear walls

• A forty-story building with double tube and a forty-story building with tube core system

• A medium-high-rise office building

(b) Frame work of design method

(c) Design criterion

(d) Literature survey for input ground motion for structural design

(e) Structural systems of foundations and their structural performance

3. Research on Reinforcing Steel Bars and Confined Concrete

3.1 Production of High Strength reinforcing bar

Several steel makers have developed three types of high

strength steel, the specified yield strength of which was

7,000 kgf/cm^ - 10,000 kgf/cm^ (700 MPa - 1,000 MPa).

The mechanical properties seemed to satisfy the criterion determined Ln the Project. The total amount of the supplied steel bars were sixteen tons until March 1990, which

were used for tests in the Project.

3.2 Pull-out Strength Test of Standard Hooks for Stirrups

Pull-out tests of seventy standards hooks which were em- bedded in high strength concrete were carried out. The

shape of standard hooks used were in accordance with

JASS5 (AIJ Standard Specification for Reinforced Concrete Construction), which is permitted to apply only to normal strength concrete. Figure 1 shows the specimens

and loading method.

Test parameters include (l) type of hook (90 deg. bent and 135 deg. bent), (2) concrete compressive strength ( 400, 800 and 1,200 kgf/cm2( 40, 80 and 120 MPa) ) and

(3)diameter of stirrups (DlO and Dl3).

From the test results it is revealed that the hooks with

yield strength of 8,000 kgf/cm^ (800 MPa) have enough

pull-out strength beyond yield stress of the steel bar, if they are embedded in high strength concrete with compressive strength more than 800 kgf/cm^ (80 MPa), using

the specification cited in JASS5.

3.3 Failure Criterion of High Strength Concrete

under Biaxial Compression

High strength concrete panels with dimension of 200 x 200 X 50 mm were loaded to failure under biaxial compression. The test conditions were simiJar to the test of

KQpfer et al. [l]. The panels were subjected to monotonically increasing load, maintaining the ratio of principal

stresses (1) 0.0, (2) 0.2, (3) 0.52, (4) 0.75 and (5) 1.0. In order to prevent the effect of confinement of loading

plate, tefron seats with cup grease were inserted between

loading plate and specimen. Stress-strain relations were measured and failure criterion are examined.

Figure 2 shows the failure criterion of the principal stress. The tests results show that the strength increases more

than thirty percent than strength under uniaxial compression, although the increase of compressive strength decrease to zero in case of the ratio of 1.0.

3.4 Development of Longitudinal Bar in Beam of

Beam Column Joint Panel

Pull out tests of 90 deg. hooked bar anchorage used for development of longitudinal bar of beam in exterior beam

column joint panel were conducted. The specimens were column with dimension of 300 x 300 x 1,500 mm and a

pair of hooked bar anchored in them. The test parameters

were (1) concrete strength (800 and 1,200 kgf/cm" (80

and 120 MPa ), (2) amount of transverse reinforcement Ln column, (3)Diameter of anchor bar, (4)thickness of cover concrete and (5) effect of multi-layer of bar anchorage.

The test results showed failure due to splitting of the cover concrete Ln all specimens.

The pull out strength observed were compared with existing Equation as (l) (see Fig. 4);

P = wdJt^^^smd- : (1) h- ]

where, w = ^x/2rcos(5r - d), 8 - tan~^(/d/,/;), Ijh = h +

r + dk, p = (r/3tii,)-0.84, = aj^, a = IS.ICqK,

7=1-1- 20Aj/{ljs), dh = diameter of anchor bar in cm,

k — distance between two contraflexure points of column

in cm, r = rudius of bent in cm, ag = concrete compressive strength in kgf/cm', /j =development length in cm, A, =sectional area of a set of transverse reinforcement Ln cm 3 — space of transverse reinforcement in cm, ; —

distance between tensUe reinforcement and concrete compressive force resultant and Cq = concrete cover thickness from surface of column to center line of steel bar. The average and standard deviation of ratio of observed

strength to calculated strength are 1.11 and 0.11, respectively.

4.1 Flexural Behavior of Column under High Axial

Stress

Experimental study on columns was performed to investigate the relationships between the amount of hoops and

deformation capacity of the column subjected to bonging moment and shear force under high axial stress. The

effects of amount and details of hoops on the deformation capacity of columns were discussed based on the test results Eight columns (250 mm x 250 mm in section and 1,000

mm in height) with loading stubs at each end of the col- umn, were designed to faU in compression after flexuraJ yielding (see Table 1). Parameters in the tests were se- lected from the factors which would affect on the deformation capacity of column. Those are 1) amount of hoops, 2) strength of hoops, 3) diameter of hoops, 4) tie reinforce- ment Ln core section, 5) arrangement of longitudinal steel bars, and 6) axial stress level (tj^. — axial force divided by

389
compressive strength of concrete and cross sectional area

of column).

Compressive strength of concrete was 1,015 kgf/cm^ ( 104 MPa ). Tensile yield strength of longitudinal steel bars were 3,870

kgf/cm- ( 395 MPa ) and 3,460 kgf/cm" ( 353 MPa ), for Dl3 and D19, respectively. Tensile yield strength of hoops were 3,500 - 1 1 ,500 kgf/cm"

( 364 - 1,173 MPa ). Bending moment and shear force were applied to the spec- imen in alternate direction with contioLLing the deformation of the specimen and keeping the loading stubs parallel to each other. Constant a.xial load the stress level r;<- of which was 0.35, was applied to the Specimens B1-B7.

Variable ajcial force, whose ajcial stress level (tj^) is from

0.0 to 0.5, was applied to the Specimen B8. Conclusions Listed bellow were obtained

1) The New RC column have a large deformation capacity

if the high strength reinforcement for hoops are properly

arranged.

2) The column with tie reinforcement in core section has

larger deformation capacity than that without tie rein- forcement. Although both of the columns have the same amount of confinement.

3) The column whose longitudinal steel bars are arranged

in the centers of the sides of the section shows larger de- formation capacity than that with longitudinal reinforce- ments only at the four corners of the section.

4.2 Flexural Behavior of Shear Wall

Static loading test was carried out on the lower part of the shear wall in the high-rise (40 stories) New RC building. The objective is to clarify the typical restoring force characteristics of high strength concrete shear wall failed

in flexure. Two shear wall Specimens, N VV-1 and NW-2, were tested. NW-1 is 150cm in span length, 20cm x 20cm column in section, 8cm in wail thickness, and 300cm in height. NW2 has the same dimension as NVV-1 except the height is 200cm. Compressive strength of concrete was 894 kgf/cm"

( 91 MPa ) for NVV-I, 955 kgf/cm^ ( 97 MPa ) for NW2, respectively. Tensile yield strength of steel bars were

7,914 kgf/cm^ ( 808 MPa ), 10,216 kgf/cm" (1,042 MPa),

and 12,846 kgf/cm" (1,311 MPa ) for longitudinal steel bars of column, reinforcement in wall panel, and hoops of column respectively. Horizontal force was applied to the top of the specimen

with the axial force of 180 tonf equivalent to the ajcial force of the assumed 40-story New RC building. Shear span

ratio were 2.0 for NVV-1 and 1.33 for NW-2, respectively.

The following conclusions are obtained

1. Longitudinal steel bars of a boundary column of NW-1 ruptured at the drift angle of 0.025 rad. Compressive failure in wall panel occurred at 0.015 rad.

in the Specimen NW-2. But columns of the speci- mens were stable and could sustain the ixiaJ force.

2. The restoring force characteristics were "S" shape

with poor energy absorption type hysteretic loops.

3. Flexurai strength of New RC shear waJl was esti- mated by the same equation for normal strength shear walls.

4. Shear strength of New RC shear wail was underes- timated by the same equation for normai strength

shear wails.

4.3 Shear Strength of beam

Objective of this study is to examine the equation to es- timate the shear strength of beams using high strength concrete of 1,200 kgf/cm" (122 MPa). The results will be

used for the future research plan to establish the equation

for shear strength of beams applicable to wide-range of concrete strength. Four specimens were tested- The specimens are 15 cm

X 30 cm in section and 90cm in span length and have

concrete stubs for loading at each side of the specimen.

Compressive strength of concrete was 1,136 kgf/cm" (116

MPa). Tensile yield strength of reinforcement was 10,160

kgf/cm^ (1037 MPa) for longitudinal steel bars (Dl6) and

2,953 - 7,993 kgf/cm^ ( 301- 816 MPa ) for shear rein- forcements (Dl6 to D8) (see Table 2). Static loading in one direction was applied to the speci- men.

Conclusions Listed bellow were obtained.

1. Empirical equation proposed by prof. T. Arakawa

underestimates the shear strength of New RC beams.

2. Shear strength of New RC beams could be estimated

accurately by the A-Method [2] proposed by the Architectural Institute of Japan using the effective compressive strength of concrete proposed by CEB

[3],

4.4 Shear Behavior of Exterior Beam Column Joint

Panel

Objective is to clarify the effect of concrete strength, amount

of hoops in joint, and axial force in column on the shear

strength and restoring force characteristics of beam col- umn joint subassembledges.

Four specimens were the 1/3 scaled exterior beam col- umn joint subassembledges of full scale frame with 3.5 m

in story height, 7.15m in bay span, 75 cm x 75 cm in sec- tion of column, and 60 cm x 75 cm in section of column.

Compressive strength of concrete was 525 kgf/cm^ ( 54 MPa ) - 907 kgf/cm" ( 93 MPa )

. Tensile yield strength

of steel bars wa^ ll,100kgf/cm- ( 1133 MPa ) for longi- tudinal steel bars (Dl3) of column and beam, and 8,400

kgf/cm- ( 857 MPa ) for longitudinal steel bars (DiO) of beam (see Table 3). The following conclusions were obtained.

1. At least 17 % of compressive strength of concrete can be expected for shear strength of exterior beam

column joint panel using concrete that is less than 800 kgf/cm^ ( 82 MPa ) in compressive strength.

2. The effect of axial force on the shear strength of New RC exterior beam column joint panel is little

if the axial stress is less than 30 % of compressive

strength of concrete.

3. Shear crack strength can be estimated by the principal stress method.

5. Research ou Structural Design

5.1 Structural Design and the Check of Their Seismic Behaviors

390
TraiaJ designs olsLxty-story moment resisting (rame struc- ture and wall-frame structure, forty-story double tube

structure and tube-core structure, and fifty-story struc- ture with long span were carried out and seismic behaviors

of the buildings were examined.

Key plan of 60 story moment resisting frame structure is illustrated in Fig. 12. Interstory height and dimensions

of structural members are tabulated in Table 4 and 5, respectively.

Span length of 6 m and building height and stance are 175.6 m {IFL = GZ -1-0.5 m) and 36 m, respectively. The

ratio of building height to building stance is 4.88. The maximum dimension is 85 cm x 85 cm for column

and 45 cm x 90 cm for beam. Material strength is listed

in Table 6. Concrete strength is from 630 to 990 kgf/cm"

( 64 - 101 MPa ), and Steel yield strength is 7,000 or 8,000

kgf/cm^ ( 714 - 816 iMPa ). The main results of design

and seismic behavior are summarized in Table 7. In the structural design using New RC materials, allowable per- manent axial stress for column, allowable interstory drift at the design base shear, shear stress of beam column joint

panel, will be important factors to dominate the building

design. However, shear stress of beams, or columns are very small, and, non-linear earthquake response analysis

showed relatively small values of the drift. Super highrise moment resisting frame structures will be ea^sily de- signed if New RC material merits are used. The following

items should be discussed to propose and establish design

method for New RC buildings.

1. The characteristics in long period range (more than

3 second) of earthquakes motion.

2. Safety for column subjected to up)-down component

as weU as two direction components of earthquake

motion.

Fifty-story ofRce buOdings with long span need structural elements such as shear wall to stiffen buUdings. One per- cent drift is commonly used as the building drift limitation against earthquake motions with the ma.ximum velocity

of 50 cm/sec. This limitation for fifteen through twentystory buUdings need to be discussed still more.

5.2 Foundation system

It is difficult to use large diameter RC pile in site, although

the foundation of New RC building is subjected to large

ajcial force and span is small comparing with that of steel buUdings.

The foundation system of three trial designed buUdings

(60-3tory moment resistant frame and waU-fiame structure, and 40-story double tube structure) are examined.

FoUowing three foundation systems are concluded to be

3uitab.'» for New RC buUdings.

1. wall type pUe

2. foundation mat with smaU diameter pUe

3. foundation using upper two systems together

CONCLUDING REMARKS

This paper described the outstanding research results mainly

concerning to structural performance conducted in second

year of five years New RC Project. The research products

to this time seems to indicate that the objectives expected

early in this program will be fruitfuUy completed. Much

cooperation and assistance to the Project is highly appreciated.

ACKNOWLEDGMENT

The Project is going to be conducted by great

effort of peoples in universities, private

companies, etc.. under the Technical

Coordinating Committee chaired by Prof. H. Aoyama. University of Tokyo.

Their works and cooperation are greatly

acknowledged.

References

[1] H. Kapfer, H. K. Hilsdorf, and H. Rusch, "Behavior

of Concrete under Biaxial Stresses, " 1969, Aug. ACI

Journal, pp. 656-666.

[2] Architectural Institute of Japan, ^Design Guideline

for Earthquake Resistant Reinforced Concrete Buildings Based on Ultimate Strength Concept "

, May

1988 (in Japanese).

[3] Comite Euro- International du Beton/Federation In- ternationale de a Precontrainte, " CEB-FIP Model

Code for Concrete Structures 3rd Edition, Paris, 1978.

391
Table 1: Lists of Column Specimens

ID Longitudinal

Re-Bars

Ratio of

Axial Stress

Hoops

Arrangement pw way pw wGy

D 1 ^ o r\ ^ fj1 U 1 d 0.35 4-5(f) @60 0.50 7892 39.5

b 2. 0.35 4-5(}) @40 0.75 7892 59.2

b J HO r» H o 0.35 4-5.5(})@60 0.61 3570 21.2

b '4 HO r\ H o 0.35 4-5(}) @60 0.50 1 1 482 57.4

1 Z - U 1 o n on; 2-5({) (5)30 0.50 7892 39.5

B6 1 2-D13 0.35 2-74) @60 0.50 8738 43.7

B7 4-D19 0.35 2-5({) @30 0.50 7892 39.5

B8 12-D13 0-0.5 4-54) @40 0.75 7892 59.2

Ratio of Axial Stress = Axial force/(bDoB)

pw : Ratio of Hoops (%) way : Yield Stress of Hoops (kgf/cm2)

SPECIMEN SPECIMEN B5.B6 SPECIMENS?

B1-B4,B8

Table 2: Lists of Beam Specimens

ID Hoops

Arrangement pw way pw way

PB-1 4-D6 (5)100 0.85 4275 36.5

PB-2 4-D8 @50 2.64 2953 78.0

PB-3 4-D6 (5)100 0.85 7993 67.9

PB-4 4-D8 (5)50 2.64 7413 1 95.7

pw : Ratio of Hoops {%)

way : Yield Stress of Hoops (kgf/cm2)

Table 3: Lists of Exterior Beam Column Joint Assemblage Specimens

ID Concrete Ratio of Joint

Strength Axial Stress Hoop

(I<g/cm2) pw (%)

Jl 525 0.0 0.60

J2 834 0.0 0.60

J3 886 0.0 0.23

J4 907 0.3 0.60

Ratio of Axial Stress = Axial force/(bDaB)

pw : Ratio of Hoops {%)

392
Table 4: Story Height

Story Story height(clear height)

60F - 41F

40F - 21F

20F - 2F

IF

2.90m (2.20m)

2.90m (2.15m)

2.90m (2.15m)

4.00m (3.10m)

Table 5: Dimensions of Member Table 6: Materials

(a) Column

Story Dimension

60F - SlF 75x75cm

50F - 41F 75xT5cm

40F - 31F 80x80cm

30F - 21F 80x80cm

20F - llF 85x85cm

lOF - IF 85x85cm

(b) Beam

S tory Dimension

RF - 52F 40x70cm

SlF - 42F 40x70cm

41F - 32F 40x75cm

3lF - 22F 40x75cm

21F - 12F 45x75cm

llF - 3F 45x75cm

IF 45x90cm

(a) Concrete

Concrete

Story specified strength

in kgf/cm"

60F - SlF Fc =- 630

50F - 41F Fc = 630

40F - 31F F^ = 810

30F - 2lF F^ = 810

20F - llF F^ 990

lOF - IF Fc = 990

(b) Steel

Diameter Specified yield

strength ay

in mm in kgf/cm^

D10(4 = 10) - D15(4 = 16) SD80(ay=8,0OO)

Dl9((ii, = 19) - D32(cf5 = 32) SD70(ay=7,000)

D35((ib = 3S) - D4l(£f6 =41) SD70(^y=7,000)

393
Table T: Main Results of Structural Design and Their Discussion

Design method Ultimate strength

design

Fc (kgf/cm2)

(kgf/cm'^) 990 7,000

Design base shear coefficient (Cg) 0.0684

Drift angle at Cg 1st, story

Max

.

1/355

1/113 (32F)

Drift angle at yield

mechanism

Base shear

1st. story

Max.

0.074^)

1/ Z JO 1/88 (33F)

Parmanent axial

stress level of column

(N/BD. kgf/cm^)

Interior

Exterion Corner

0.249FC

0.152FC

Max. steel ratio in column

Interior

Exterion

Corner

1.32% 2.38% 3.79%

Max. tensile steel

ratio in beam

1.39%

Shear stress of column

(Q/BD, kgf/cm^)

Static

Dynamic^^

18.6

14.9

Shear stress of beam

(Q/BD, kgf/cm^)

Static

Dynamic •' 23.2 Ductility factor-0.7

Additional axial

force of column

( AN/BDFc)

Static

Dynamic^^

0.511Nu^) (Corner)

0.188 (Taft EW)

Dynamic response^

Drift of IF Max. R

Ground motion

1/716

1/227 (45F) 0.040

Taft EW 1952

Dynamic Response^^ Drift of IF Max. R Cb

Ground motion

1/444

1/186 C25F1 0.0541 Taft EW 1952

Natural period (sec.) 1st. 2nd.

3rd.

4.02

1.28 0.69

Max. shear stress

and bond stress in joint panel (kgf/cra )

Shear stress

Bond stress

105.5 (3.35 /^)g.

less than 4 JTc '

1) Max. velocity of input ground motion is modified to 25 cm/sec.

2) Max. velocity of input ground motion is modified to 50

cm/ sec

.

3) The second reflection point on the diagram of shear v.s. deformation relationship.

4) Nu - 0.85(B*D-Ag)Fc+Ag <J y Ag : total amount of steel in column

5) Bond stress is calculated as 1.3(1.1 y)d/(4Dc)

d : diameter of steel

Dc: depth of column

394
f 1 direction of concrete cure

I 30

30

! 300

(400)

-010 (D13) 0

load ceil(tension)

/ chack

022

30

c

crO

a. U-3

_! 200

DIO 0

ring

oil jack specimen

strain gauge displacement

transducer load cell(compression)

Figure 1: Loading Apparatus for PuU Out Test of Standard hooked Anchorage

395
2b

Na6

20

/

/--^

iNaS

co 15

• ' /

Na 4 ; 500

cu 10 Na5 ; 984

5

. Na6 ; 1061

0 1 2 3 4 5

S ( mm )

(a) Effect of Concrete Strength

iNa 4

25 --^-^N

20 /• /

Na5

/ / co•4-* 15

If

1

1

10

1/ ' Na5 ; r=3db

5

^ Na 9 ; r=5db

0 1 2 3 4 5 6

Na9 (N)

1

1

1 1

1

S ( mm )

(b) Effect of bent radius Figure 3: Pullout Load - Slip Relations

of 90 deg. hooked anchor

P=P--h/{h-i)

F= Pd- sin d

Figure 4: Schematic Mechanism of anchorage

of 90 deg. hooked anchor

c

o

UJO

cco

LL<

z

o

N

tro

X

40

20

0

-20

-40

40

20

0

-20

-40 :±;iB6 : 87

1 i-iB4

-40 -20 0 20 40 -40 -20 0 20 40 -40 -20 0 20 40

DISPLACEMENT (mm)

Figure 5: Horizontal Force and Displacement Relationships

396
160

120

80

40

0

-40

-80

-120

160

Shear Crack

Flexural Crack ~)

> 1 1 •—

rColumn Re-Bar Yield V- - ^Wall Pe-Bar Yield

/i^^\j>;^y''^y^ '

• -

-HW^ ~ -—-^^^ i-rrr-T. W \

\ i

i . i . Specimen NW-1

30 -20 -10 0 10 20 30

Rotation Angle (x10-3rad.)

^Wall Re-Bar Yield Z ^ -J

160

-30 -20 -10 0 10 20 30

Rotation Angle (x10-3rad.)

Figure 6: Shear Force vs. Rotation angle Relationship

Figure 7: Crack Pattern

397
PB-4 (Flexural Yielding

and Bond Splitting)

Deflection (cm)

Figure 8; Shear Force vs. Deflection Relationships

30 180 5 100

"T

13 151513

25 31 32 31 25

III II ii II I

-125

25 100100 25

- 1192

1600

25 100 100 25

J—278

[

Figure 9: Specimen

398
-80 -40

(x10-3radian)

40 80

0 Story Drift Angle

/' -8 ^- Maimum strength

=

J2

Figure 10; Skeleton Curves of Shear Force of column vs. Story Drift Angle

CM S E

o o 160

" ll20

80

V) 1_>

CO "

CO

TO<u 40

sz /

I

J1

J3

fi J4

J2

^ Positive

<^ Negative

Loading

Joint Shear Failure

200 400 600 800 1000

Concrete Strengtti ( OB kg/cm2)

Figure 11: Relationship of Maximum Shear Stress vs. Concrete Strength

6.00

6.00

6.00

36.00

6.00

5.00

5.00 -« -

(Unit : m) 6.00 I 6.00 I 6.00 , 6. 00 , 6.00 , 6.00

36.00

Figure 12: Plan

399
Characteristics of Supplemental Dampers for Earthquakes

by

Robert D. Hanson^

ABSTRACT

A number of imaginative approaches to improved

earthquake response performance and damage

control have been developed and others will be

forthcoming. These can be divided into two groups.

Passive Systems of which base isolation and

supplemental mechanical damping are examples and

Active Systems which require active participation of

mechanical devices whose inputs depend upon

measured building response. The discussion in this paper focuses on only passive supplemental

mechanical damping systems. It is hoped that as a

result of the information summarized in this paper

the reader will be able to conclude that supplemental damping can provide performance and

economical advantages for the earthquake resistant design for some new buildings and provides other

advantages when developing strengthening systems

for existing buildings. The advantages and

disadvantages among the damping devices discussed

is left to the reader. However, information

necessary to make decisions between the use of

supplemental damping systems, base isolation or

active control systems is beyond the scope of this paper. KEYWORDS: Damping, Damping devices.

Structural response, Supplemental damping

1. INTRODUCTION

The building motions and damage caused by the

Chilean and Mexican earthquakes of 1985 have

amply demonstrated that single measures of

earthquake motions, such as peak ground

acceleration, do not provide an adequate basis from

which to anticipate building response amplitudes or

building damage. Even the use of multiple measures of ground motion as incorporated in the

Response Spectra do not explain the survival or

failure of buildings subjected to different earthquake ground motions. Many engineers

believe that the duration of the ground motion is an important factor in exciting the response and

causing damage to a building. Some have suggested

that an additional design parameter is need to complement the current Response Spectral

procedure (Code Lateral Force procedure).

Earthquake input energy is currently being

investigated as a logical complementary

consideration for the design of earthquake resistant buildings.

Simultaneously, the expectations of building owners and engineers are increasing. That is, they expect

that their buildings will not experience serious

damage in a large earthquake, such as occurred in Mexico City in 1985, and that the buildings will continue to be operational following moderate

earthquake ground motions, such as the Whittier

event in 1987 and in San Francisco during the Loma Prieta earthquake of 1989. Limits on lateral

drift calculated for code seismic forces are being

used implicitly to control building damage. These

greater expectations for the behavior of our

earthquake resistance building designs should be

faced directly with knowledge and imagination

rather than indirectly through implicit methods.

The paper is divided into two main sections. The

first summarizes analytical studies of the earthquake

response of buildings with added supplemental

damping in both the elastic and inelastic range of response. The results are presented in a form of

response spectra modifications which are appropriate as damping is added to the building.

Although data also exists for presentation as energy

input spectra modifications, that data is not included in this paper. The second section of the paper discusses the characteristics of the two most

practical classes of supplemental energy dissipation

devices for earthquake resistant design utilization and illustrates some of the commercially available

devices. Experimentally determined characteristics

for several of these devices are given. It is

^National Science Foundation, Washington, DC

20550

400
concluded that added damping can provide

significantly improved earthquake response

performance and that practical commercial damping

devices are available.

2. EFFECT OF ADDED DAMPING ON

SEISMIC RESPONSE

The displacement response spectra, SD, is a key

parameter in estimating the maximum displacement

responses in each mode of a building for past

earthquakes or for a specified design spectra. This

spectral displacement decreases as damping

increases. Ashour (1987) developed a relationship

for the change in SD of elastic system with changes

in damping and correlated these with results obtained from existing earthquake accelerogram

records.

2.1 Elastic Response.

The natural period, Tn, used in the study were 0.5,

1.0, 1.5, 2.0, 2.5, and 3.0 seconds which covers a

representative range of natural periods. Three real and twelve artificial earthquake records were used

for excitation input. Damping values used were 0,

2, 5, 10, 20, 30, 50, 75, 100, 125 and 150 percent of

critical. The values of SD for a given Tn and

damping factor were normalized with respect to the

SD at Tn for zero damping for each earthquake

record and were than averaged over the 15 records

to obtain a mean value for each period and fraction

of critical damping. Figure 1 shows the resulting

relation for the mean value of SD as a function of

period and damping for zero and five percent

damping normalizations.

These curves can be represented by simple decaying

functions,

R = [{1 - e-fi^}/{fiB}]

where ^ is the selected fraction of critical damping

and B is a coefficient which was evaluated for zero initial damping normalization to be 24 for the

upper bound and 140 for the lower bound. Figure

2(a). For an initial elastic spectral normalization of

a, the simple decaying function can be expressed by

R = [a {1 - e-^^}/^{l - e-<^}y'^

The system damping must be greater than or equal

to the initial elastic spectral normalization damping

to use this equation. For five percent (5%)

damped spectra normalization, set a = 0.05. Ashour (1987) found the upper bound B = 18 and

the lower bound B = 65 for this case. Figure 2(b).

2.2 Inelastic Response.

The inelastic response evaluations are more difficult to establish on a comparative basis. Wu (1987)

used one artificial and nine real earthquake records

to study the elastic-plastic response of single degree

systems. One form for presentation of his conclusions uses the peak ground parameters as a basis to derive the inelastic response spectra.

Figure 3 gives the Response Spectra amplification

factor in the acceleration region at periods of 0.1 second and 0.5 second for damping from 10 to 50

percent and ductilities from one to six. The effect of increasing damping and ductility can be seen. The amplification factor in the velocity region is shown in Figure 4 and in the displacement region

at periods of three seconds and ten seconds in Figure 5. In these figures $ is the fraction of

critical damping and n is the structural ductility

factor. Thus, it can be seen that supplemental

damping can effectively reduce structural yielding

demands of an earthquake.

It is interesting to note that these amplification

factors can be divided into terms which include the viscous damping of the system and separate terms which include the structural inelastic yielding. The

inelastic deamplification factors for the three

regions of concern are plotted in Figure 6 with the

corresponding data including damping from 10 to 50 percent of critical. The relatively small scatter of the data with changes in damping illustrates the point that spectral modifications for high damping

and for inelastic action can be considered

separately. Thus previous experience with inelastic spectral modifications can be retained while

incorporating modifications for higher damping.

As will be seen later steel yielding hysteretic

characteristics are generally curvilinear and can be

reasonably represented by equivalent viscous

damping as was assumed in the previous studies. Friction and lead yielding characteristics are box

shaped with sharp corners; for large motions

equivalent viscous damping which changes with

amplitude can be used, but for small motions the behavior is more of a stick-slip nature for which

equivalent linear modelling is not very accurate. For these reasons most of the friction (and lead

yielding) analytical studies have used nonlinear

401
models to represent the damping device

characteristics.

Filiatrault (1988) described a design procedure for simple friction damped braced frames. The system

used the Pall friction damping device located at midpoints of diagonal braces, Figure 7. They

assumed that the key parameters in the design process were the expected peak ground acceleration,

a^ the predominate period of the ground motion,

T^ the period of the unbraced building, T^; the

period of the braced building (no slip of the

devices), T^; the mass of the building, m = W/g;

and the friction slip load, V^. Using a stochastic

representation of the ground motion in a parametric study they develop the simple friction

slip load design spectrum illustrated in Figure 8.

Given the weight of the building, its fundamental

unbraced period, the expected braced frame period

(with no slip), the expected peak ground

acceleration, and the expected predominate period

of the ground; the design friction slip load is determined using the curves of Figure 8. There are

several parameters which can greatly influence the

selected slip load. First, the design slip load is selected to be directly proportional to the expected

peak ground acceleration and the weight of the

building. Second, the ratio of the ground motion

period to the unbraced frame period has

significantly different characteristics for different period ratios: For T^ less than T„ the design slip

load changes rapidly with T^. For T^ greater than

Ty the design slip load is relatively independent of

T„, but is dominated by the ratio of the braced

period to the unbraced period.

Studies at the University of British Columbia on

the Pall friction damper have used the minimum

elastic energy of the structure (sum of

instantaneous strain and kinetic energy of the

structure) as a key parameter in setting the

optimum slip load for the friction devices. Using

the first six seconds of the El Centre 1940 NS

accelerogram a curve showing the variation in maximum system elastic energy as a function of slip

load is given in Figure 9. It can be seen that the

system energy is sensitive to the selected optimum

slip load. At zero slip load the structure behaves

as an unbraced frame and at very high slip loads it behaves as a concentrically braced frame. Thus, the

selection of the appropriate slip load must consider

both the expected moderate earthquake motions for damage control and the potential severe earthquake

motions for safety.

3. DAMPING DEVICES

Damping devices can be classified into four primary

types on the basis of the material used to transform

mechanical energy to heat: (i) Velocity proportional

viscous material which can be liquid, such as silicon

oil, or solid, such as special rubbers or acrylics; (ii) Friction devices in which the resisting forces are a constant which depends upon the interface contact

pressures and the interface materials. Steel to steel, copper with graphite to steel, or brake pad to steel interface materials are most common; (iii) Metallic

yielding devices in which the resisting forces depend

upon the nonlinear stress-strain characteristics and

geometrical configuration of the material. Mild

steel and lead are the two most commonly used

materials; and (iv) Magnetic damping devices. The

following discussion will concentrate on the friction and metallic yielding devices because they appear to be more practical for earthquake resistance

applications.

3.1. Friction Devices

3.1.1 Cylindrical Device (Sumitomo Metal

Industries) The standard device (Figure 10) has a 10 ton (22 kips) maximum slip capacity and an 8

cm (3 inch) double amplitude stroke range (Figure

11). This device saw its initial application as the

shock absorber for the Japanese "bullet" train. The

potential use of this device in a 31 story building was reported by Teramoto (1988).

3.1.2. Pall Friction Braced Frame Device (Pall

Dynamics) The direct slip device uses brake pad

material against steel. The normal pressure which

controls the slip force is maintained by a spherical

springs. The device is located in the structure as

illustrated in Figure 7. The characteristics of the

device have been reported earlier (Pall, 1982; Pall, 1986). It was reported that the device has been

installed in a building in Canada.

3.1.3 Rotary Friction Device (Hazama-Gumi) The

device uses sintered metal brake pads against

stainless steel to provide the friction force. Because

the rotational characteristics of the device the

direction of the input forces causes a slight

asymmetry of the frictional resistance (Takai, 1988).

This device was developed for use with base

isolation systems, but could be adapted for in structure use.

402
3.2. Metallic Devices

3.2.1 Lead Damper (DSIR-New Zealand, Giles

Corporation, (Fujita, 1988)) The piston lead damper (Figure 12) was developed in New Zealand

for bridge uses and has been licensed in Japan by

Giles Corporation. A different lead damper uses a lead post containing 12 steel wires to ensure pure

shear deformation in the lead (Figure 13). Sample

data shown in Figure 14 illustrates that the lead

hysteretic characteristic is very similar to the

friction damper characteristic.

3.2.2 Steel Plate Damper (Bechtel Corp.,

CounterQuake Corp.) The device was originally

developed for pipe restraints in nuclear power

plants. The concepts were extended to building

structures (Scholl, 1987). The configuration of the

device (Figure 15) is selected to maximize the

amount of material yielding without excessive

localized strains. Experimental data on smaller

sized ADAS devices have been reported (Bergman,

1987; Bergman, 1988) and are illustrated in Figure

16. Three story steel frame shaking table tests using these devices has been recently completed at the University of California at Berkeley (Whitaker,

1989).

3.2.4 Steel Post Dampers (Kajima Corp., Giles

Corp.) The basic concept of these devices uses

cantilever beam yielding. The dimensions and

cross-sectional shape along the height are unique.

The Joint Damper (JD) by Kajima is used to connect separated structures at the atrium roof

level. The connection provides stiffness and energy

dissipation characteristics. The Giles Steel Bar

Damper is used to provide vibrational isolation and

hysteretic damping during excessive motions.

4. REMARKS

Analytical studies have demonstrated that supplemental damping can effectively control the

response of buildings during earthquakes. These

studies have shown that the response of a highly

damped building is not sensitive to the peculiarities

of specific earthquakes. Thus the added viscous

type damping effectively decreases response

variability and uncertainty from a broad spectrum of

earthquake inputs; thereby improving the

performance of the building and reducing the risk of severe damage.

A large number of damping devices are currently

available and others under development. This is a tribute to the imagination of earthquake engineers.

While most of the devices developed in Japan are intended for use with base isolation systems for the

nuclear power industry, application to commercial

and residential buildings is moving rapidly.

5. CGNCLUSIGNS

The following conclusions are the result of a number of analytical studies and the information

summarized herein:

5.1. The incremental effect of adding supplemental

damping devices is greater when the total damping

is small than when the total damping is high.

5.2. The combined effect of added damping and

inelastic response of the structure can be separated

into the effects resulting from higher damping and

the effects resulting from member yielding.

5.3. Supplemental damping devices are more

effective in reducing the corresponding earthquake

response spectral values in the mid period region

(velocity region) than in the low (acceleration

controlled region) and high (displacement

controlled region) period regions. Nevertheless

additional damping makes the structures less sensitive to earthquake peculiarities.

5.4. The selected damping devices should be

dependable and require a minimum amount of

maintenance. The delivered devices should have

quality controlled manufacturing to ensure that the

characteristics established through testing programs

and specified by the designers is achieved in field installations.

6. ACKNGWLEDGEMENTS

This paper is based upon research supported by

National Science Foundation Grants ECE-85 12726

and CES-8821735, and by the Bechtel Power

Corporation. The opinions expressed herein are those of the author and do not necessarily

represent the opinions of the sponsors. Much of

the analytical and experimental research

summarized was obtained by Samir Ashour

(Ashour, 1987), David M. Bergman, Yung-feng Su

(Su, 1990) and Jian-ping Wu (Wu, 1987) in conjunction with their Ph.D. studies at the

University of Michigan. Their originality and

contribution to this effort are gratefully

403
acknowledged. Portions of this paper were

prepared for the 1988 SEAOC Annual Convention

(Hanson, 1988).

7. REFERENCES

Ashour, Samir A., (1987) Elastic Seismic Response

of Buildings with Supplemental Damping, Ph.D.

Dissertation, Department of Civil Engineering,

University of Michigan, January.

Bergman, David M. and Subhash C. Goel, (1987)

Experimental Characteristics of the ADAS Device,

Department of Civil Engineering, University of

Michigan, Report UMCE 87-10, November.

Bergman, David M. and Robert D. Hanson, (1988)

Characteristics of Mechanical Dampers, Ninth

World Conference on Earthquake Engineering, Tokyo/Kyoto, Japan, August, (#6052).

Filiatrault, Andre and Sheldon Cherry, (1988)

Seismic Design of Friction Damped Braced Frames,

Ninth World Conference on Earthquake

Engineering, Tokyo/Kyoto, Japan, August, (#5145).

Fujita, Satoshi, Takafumi Fujita, Yukio Sasaki,

Shigeru Fujimoto, Noboru Narikawa and Chiaki

Tsuruya, (1988) Earthquake Isolation Systems for Buildings of Industrial Facilities Using Various

Types of Damper, Ninth World Conference on

Earthquake Engineering, Tokyo/Kyoto, Japan,

August.

Hanson, R.D., D. M. Bergman and S.A, Ashour,

(1986) Supplemental Mechanical Damping for Improved Seismic Response of Buildings, Third

USA National Conference on Earthquake

Engineering. Charleston, SC, August, Vol. II, pp. 1129-1140.

Hanson, Robert D., (1988) Energy Dissipation

Systems, Proceedings of the 1988 Structural

Engineers Association of California Convention, Kona, Hawaii, October

Pall, AS. and C. Marsh, (1982) Response of

Friction Damped Braced Frames, Journal of the

Structural Division. ASCE, Vol. 108, No. 6, June,

pp. 1313-1323.

Pall, A.S., (1986) Energy-Dissipation Devices for Aseismic Design of Buildings, Proceedings. ATC

Seminar on Base Isolation and Passive Energy

Dissipation . Applied Technology Council, Redwood

City, CA, March, pp. 241-250.

Scholl, Roger E., (1988) Added Damping and

Stiffness Elements for Earthquake Damage and

Loss Control, Proceedings of Conference XLI: A

Review of Earthquake Research Applications in the

National Earthquake Hazards Reduction Program:

1977-1987. U.S. Geological Survey, Open File Report No. 88-13-A, San Diego.

Su, Yung-feng, (1990) Seismic Response of Building

Structures with Mechanical Damping Devices, Ph.D.

Dissertation, Department of Civil Engineering,

University of Michigan, Ann Arbor, January.

Takai, Hidehiro, Tetsuo Uno and Kiyoshi Nagai,

(1988) Study on the Friction Damper as the Device

of a Base Isolation System, Ninth World

Conference on Earthquake Engineering, Tokyo/Kyoto, Japan

Teramoto, Takayuki, Haruyuki Kitamura, Kenji

Araki and Kiichi Takada, (1988) Application of

Friction Damper to Highrise Buildings, 13th

Congress, International Association for Bridge and

Structural Engineering, Helsinki, Finland, June,

pp.753-758.

Whittaker, AS., V.V. Bertero, L.J. AJonso, and C. Thompson, (1989) Earthquake Simulator Testing of

Steel Plate Added Damping and Stiffness Elements,

Report No. UCB/EERC-89/02, Earthquake

Engineering Research Center, University of

California, Berkeley, January.

Wu, Jian-ping, (1987) Statistical Study of the

Inelastic Response of Structures with High

Damping Subjected to Earthquakes, Ph.D.

Dissertation, Department of Civil Engineering,

University of Michigan, Ann Arbor, November.

Wu, Jian-ping, and Robert D. Hanson, (1989)

Inelastic Response Spectra with High Damping",

Journal of the Structural Division, ASCE, Vol. 115,

No. 7, July, pp. 1675-16%.

404
I .2 .3 .4 .5 Friction of Critical Otmping

(a) Zero Damping Normalization

.1 .2 .3 .4 .5 Fraction of Criticti OMping

(b) Five Percent Damping Normalization

Figure 1. Normalized Mean Displacement for Spectra of 15 Eanhquake Records

[from Ashour (1987)]

405
.2 ^4 TT

Fraction of Critical Damping

(a) Zero Damping Normalization

(b) Five Percent Damping Normalization

Figure 2. Envelope of Normalized Mean Displacement Response

[from Ashour (1987)]

406
i

i I

I I M

•r- O MS In (0.0959 (T) (2 89 n 1.89)- " 2**

10

B-0.10

• 6-0.20 6-0.30

* 6-0.50 — by formula

10

Ductility \L

• 0.547 in (0.417 ^ (1.82 u 0 82)- 0.562

I I I I I

6-0.10

• 6-020

6-0.30

• 6-0.50 — by formula

Ductility \i

(a) At 0.1 sec Period (b) At 0.5 sec Period

Figure 3. Amplification Factors in tlie Acceleration Region

[from Wu (1987)]

10

• 0.471 M (0.524 fl)(L53y- 0.53)- O-'OS -

J 1 I I

O

a

u.

6-0.10

• 8-020

6-0.30

* 6-0.50 — by formula

Ductility )i

Figure 4. Amplification Factors for PSV^ in the Velocity Region

[from Wu (1987)]

407
r- . 0.478 In (0.475 ^5 <i • l-0«

6-0.10

• 8-OiO

B-0.30

* 0-0.50 — by formula

10

u

a

U.ci 1 a

u

a.E

<

V = - 0 291 M (0.0473 ^ ^i- 1 06

1

1 1—1—4

B-0.10 B-020

B-0.30 8-0.50

by formula

Ductility \i Ductility p.

(a) At 3.0 sec Period (b) At 10.0 sec Period

Figure 5. Amplification Factors in the Displacement Region

[from Wu (1987)]

OuctilUy (1 Ductility p.

(a) In the Acceleration Region (b) In the Velocity Region

Figure 6. Deamplification Factors in the Three Regions

[from Wu (1987)1

408
u

(O u.c

o

a

o

Q.

E

(00) a

0-.1 at T-3 s

Q-.1 atT-IOS

B-.2 at T-3 s 0-.2 at T-10 s 6-.3 at T-3 s 6-.3 at T-IOs

C-.S at T-3 s 6-.5 at T-IOs

by lornuita

Column

Ductility n

(c) In the Displacement Region

Figure 6. Deamplification Faaors in the Three Regions

[from Wu (1987)]

Figure 7. Pall Friction Device Installation [from Filiatrault (1988)]

Figure 8. Design Slip Load Spectrum for Single Degree System

[from Filiatrault (1988)]

409
Figure 9. Optimum Slip Load for Friaion Damped Braced Frames

Subjeaed to Six Seconds of El Centro 1940 Accelerogram

[adopted from Univ. of British Columbia]

Cotter outar wadge

(splittad into 3 btoetal

Outer cylinder

spnntg

Cotter outer wedge

(splitted into 3 blocks)

At-Ai

Cro»-«ection«l vtcw

(Note) Sizes given above are for a frictioful force of 10tf.

Figure 10. Cylindrical Friction Damper

410
Oap«ndancY on Amplituda of Frictlonal Force

y > — F -

1 Spocimon ^ Jl M ft R . T PortphorsI temperature

.a. K

Wave form Sine wave V A » 0 33 HZ Frequency

Figure 11. Hysteresis Characteristics for Cylindrical Friction Damper

Bcahng LMd Inner tube

Figure 12. Piston Lead Damper

Figure 13. Cylindrical Lead Damper with Steel Wires

m «

Amplitude

20

Temperature of damper

58 r

62

60

64

0«fUslon ( Full •Urn ) X10*«

-10 -6 0 6 10 -10 -5 0

I

20

t 10

v^-10

"--20

2-30

, 1 I

^^^^

m

1 i i 1 , -3-7-1 0 12 3 -3-2-1

• I) X10

0 12 3

m

Figure 14. Hysteresis Characteristics for Lead Damper Shown in Figure 13 [from Fujita (1988)]

411
Figure 15. Added Damping and Stiffness (ADAS) Steel Plate Damper

[from Bergman (1987)]

DZSPZJLCI:KEK7~ (inches) OZSPLACEMEMT (inches)

(a) TestXTA (b) Test XATA

Figure 16. Characteristic Hysteresis Curves for Seven Plate, Five Inch High,

5/16 Inch Thick ADAS Device [from Bergman (1988)]

412
An Economic Analysis of Laminated Rubber and

Steel Bearing Base Isolation System for Buildings

by

-

Tingley K. Lew'

ABSTRACT

The economics of laminated rubber and steel bearing base

isolation system for buildings is determined from the very

limited data available. The incremental costs for installing the base isolation system for a four-story concrete frame and

six-story steel frame building are compared with the benefits from reduction in structural and non-structural costs, and the present worth of reduction in expected future earthquake

damage over the service life of the base isolated building.

Based on the results of the analysis, conclusions and recommendations are given.

KEYWORDS; Base isolation system; earthquake damage;

incremental cost; isolation bearings; percent cost.

1. INTRODUCTION

The lateral force resisting systems of conventional fixed- base buildings are generally designed to resist seismic forces by having the necessary strength and ductility. Due to the

relative large seismic design forces especially in zones of

high seismic activity, it is generally not economical nor

practical to design the lateral force resisting systems of

buildings to respond elastically under extreme earthquake

loading. Hence, the lateral force resisting systems of conventionally designed buildings are expected to respond

inelastically under the maximum earthquake-induced forces during their service lives.

Such inelastic deformations of the building's lateral force

resisting system generally produce significant structural damage. In addition, large interstory displacements (story

drift) accompanying such deformations cau.se nonstructural damage in the building. Dynamic response of the building's

structural system generally amplifies the earthquake ground

motions at the upper levels of the building. Such dynamic

amplification of the ground motions causes nonstructural damage unless adequate seismic anchorage and supports are provided. Nonstructural damage include the overturning and sliding of the building contents and subsequent impact

of the contents (for example, bookshelves, .sensitive mechanical and electronic equipment, filing cabinets, etc.).

The intent of any base isolation system is to reduce the earthquake-induced accelerations at the various levels of the building and, hence, the base shear demand on the building

(Figure 1) by increasing (shifting) the fundamental period and damping of the fixed-base building, adjusting and/or

controlling the response mass, or some other means. Historical development of the base isolation system for build- ings and its implementation are given in References 1 through 6. Some of the more recent analytical and experi- mental studies on base isolation for building structures may

be found in References 7 through 12. The friction pendulum

system (Ref 9) appears to be most promising. This simple

system allows the engineer/designer relatively fine control over the period shift of the fix-based building and is rather

insensitive to torsional response of buildings with asymmetrical ma.ss and/or stiffness distribution compiu-ed to the other base isolation systems. Base isolated building structures in Japan are given in Reference 13.

In current U.S. practice, base isolation is generally accomplished by introducing lateral fiexibility by way of isolation bearings/energy dissipators at the base of the building,

thereby shifting the fundamental period of the building away

from the range where the earthquake ground shaking energy

is dominant and accompanied by large demands (Figure 1). Thus, the dynamic response of the ba.se isolated building is greatly reduced as compared to the fixed-base building.

Base isolation is most suitable for structures located rela- tively close to the potential earthquake sources in high

seismicity areas with large seismic design forces. Further- more, the suitability of base isolation depends on the in- tended occupancy of the building. Certain occupancy demands an increase in safety above those obtainable from

conventional design, e.g., hospitals, fire stations, police

stations, telephone exchanges, telecommunication and computer centers^ housing es,sential services, and/or expensive equipment.

Structures with low fundamental periods «0.7 to 1.0 sec- ond) benefit the most from the increase in period by base

isolation. Generally, most of the earthquake ground shaking energy input to the base of the buildings on stiff to intermediate soil sites is concentrated in periods ranging between

0.1 and 1.0 second. Base isolation is generally best suited

for relatively stiff, squat, low- to medium-rise buildings with

'Naval Civil Engineering Laboratory Code L5I, Port Hueneme, CA 93043-5003

^The cost of equipment housed in such facilities can be up to several times the cost of the building. By contrast, the cost of the contents of a typical commercial building is about 10% of the cost of the building.

413
between three to ten stories and fixed-base fundamental

periods less than about 0.7 to 1.0 second. Base isolation is not likely to benefit taller or more flexible buildings whose

fixed base period is in the range where the demand from the earthquake site response spectrum is relatively low and

decreases relatively slowly with increasing penods (the right most portion of the curve in Figure 1).

Earthquake base isolation is best suited for buildings located on stiff to intermediate stiff soil sites. Sites with soft soil profiles tend to filter out the high frequency components of

the eanhquake ground motion and may provide some reduction of the seismic forces transmitted to short period structures. Relatively deep (deeper than about 5 feet or 1.5m)

soft soil deposits at the site can amplify the relatively long

period (> 0.30 second) components of ground shaking from

nearby earthquakes and/or large magnitude distant earth- quakes and defeat the intent of base isolation (e.g., acceleration data recorded in soft clay sites in the San Francisco Bay

area dunng the 1989 Loma Pneta northern California earth- quake). By contrast, sites with stiff soil profiles tend to amplify the low period components of the ground motion. A base isolation system on such a site will filter out these low period components of eanhquake ground shaking energy and be more effective than on a soft soil site.

The current application of base i.solation to buildings is made difficult by the limited data on the range expected

earthquake ground motion characteristics for a particular site relatively close to the energy source and the uncertainties about the inelastic structural response under relatively large loads from a major eanhquake. Thus, most of the existing base isolated buildings are designed to remain es.sentially elastic with a small amount of localized yielding in noncritical structural framing members for safety and serviceability.

To reduce the contribution from their higher modes, base

isolated buildings are generally designed to be relatively stiff

in the lateral directions. The principal dynamic response

characteristics of a base isolated building are illustrated by

the mode shapes and modal shear distribution shown in Figure 2 for a base isolated shear beam type building.

• The relatively high flexibility and damping of the base

isolated building results in a long first mode period and,

thus, low demand (loading) from the site respon.se spectrum.

• The relatively high lateral flexibility of the base iso- lated building system ensures a low modal panicipation

factor for the second and higher modes. This reduces

the horizontal accelerations and seismic forces on the building and its contents.

• The energy dissipation capacity of the base isolation system reduces the seismic forces, the associate base

isolator displacement, and building deformations.

• Ba.se isolation provides a more uniform seismic shear

force distnbution over the lower levels of the building.

Although base isolation reduces the horizontal accelerations of the building, it Increases the horizontal displacement of

the building system as a whole. The story drift of the base

isolated building is reduced. However, the horizontal displacement of the building over its height can be relatively

large, with most of the displacement concentrated at the base

of the building. The base isolation system and the utility connections at the isolator level must be designed to ac- commodate the anticipated base isolator displacement

(possibly as large as 12 to 16 inches for some systems) and

remain functional.

Due to the lateral flexibility introduced by the base isolation system, wind loading becomes one of the dominant factors

in the design of a base isolation system. Theoretically, a base isolation system can be designed to reduce the maximum floor accelerations to one-tenth or less than the maxi- mum horizontal eanhquake ground acceleration at the site. In practice, this reduction is limited by the lateral displace- ment capacity of the base isolation bearings, the wind

loading at the site, and the amount of hysteretic energy

dissipation required by the base isolation system. The

laminated rubber and steel isolation bearings must be de- signed to remain stable in the vertical direction under the maximum earthquake-induced horizontal ground displace- ment at the site. This indirectly controls the lateral stiffness of the bearings and, hence, the seismic load transmitted to the structure. Wind restrain is needed for the base isolation system to minimize personnel discomfort from the relatively

frequent wind-induced displacements of the building and the resulting fatigue damage to piping and other utility connections in the building. For high seismicity sites, the base

isolation system is generally designed to yield at between

3% and 6% of the weight of the building to satisfy wind

restrain requirements. Although some researchers have

suggested that 5% is optimal, the three base isolated build- ings that use a laminated rubber and steel bearing with a

vertical cylindrical lead plug energy dissipator in Reference 4 through 6 have a design yield of about 7% of the weight of

the isolated structure.

To provide an adequate amount of energy dissipation capacity, the hysteretic energy dissipators of the base isolation system must have a sufficiently high yield capacity to provide the needed energy dissipation.' A highly ductile energy dissipator with a low design yield level is unlikely to

'The energy dissipation capacity of the hysteretic energy

dissipator is proportional to the area under its force dis- placement curve. Ductility is the ratio of the maximum

displacement to the displacement at yield.

414
provide the needed energy dissipation to limit the maximum

horizonta] displacement of the isolation bearings to reasonable values. The effective combined lateral stiffness of the laminated rubber and steel bearing and the yielding lead plug energy dissipator controls the seismic forces transmitted to the building and the maximum horizontal displace- ment of the building. A 2-inch-thick shell of rubber covers

the outside of the isolation bearing (Figure 3) to provide about 3 hours of needed fire protection. This layer of rubber

also protects the steel plates of the bearing from atmospheric

corrosion. A properly designed base isolation system will generally result in an equivalent base shear of between 0.2g and 0.25g for the building with a maximum horizontal design earthquake acceleration of about 0.5g.

For a fixed-base building with an effective period of less than about 0.7 second, the input maximum eanhquake

ground accelerations at the site are amplified at the upper

floor levels by average* factors of about 3.5 and 2.5, respectively, for 2% and 5% damping. By contrast, both analytical

results (Ref 14 and 15) and experimental results for a onethird scale building model (Ref 16) have shown that a properly designed base isolation system can generally reduce

the dynamic accelerations of the building and, hence, base

shear demand on the building by a factor of about five, as compared to the fixed-base building. A three-bay, five- story, one-third scale model steel frame building, with and

without laminated rubber and steel/lead plug isolation bearing/energy dissipation system was tested at Berkeley

(Ref 16). The model was subjected to the 1940 El Centro

earthquake time-history, scaled to a peak value of about

0.55g. The experimental results are shown in Figure 4. From the curves in the figure, it can be seen that the maximum table acceleration for the fixed-base building is amplified by a factor of about 5 at the fifth floor level. The

average acceleration for all the floor levels is about l.25g.

By contrast, all the floor accelerations for the isolated build- ing are only about 0.25g, one-half of the maximum shake

table acceleration. Consequently, potential earthquake

damage to structural and nonstructural elements in the base

isolated building is expected to be much less than that of the fixed-base building.

Results of analytical studies, laboratory experiments on scale models of steel frame buildings on a shake table, and field performance of prototype buildings' indicate that the usage

of laminated rubber and steel/lead plug isolators for earth- quake base isolation is technically feasible. The only question that remains is whether earthquake base isolation for buildings is cost effective. For this economics analysis, the base isolated building is divided into three subsystems

(Figure 5):

• Building System

Structural frame - Bottom floor slab

Isolation bearing connections

• Base Isolation System

Isolation bearings/energy dissipators

Bearing connections to foundation Foundation

Water/gas/electrical connections

• Soil System

Soil profile and properties .

" Earthquake ground motion input

In the building system, the bottom floor slab and isolation becuing connections are additional cost items as compared to a conventional fixed-base building.

In the base isolation system, the isolation bearings, the bearing connections to the foundation, the additional excavation required to provide the gap around the perimeter of the building to accommodate building movement at its ba.se and

sliding gnlls to cover over this gap, and the retaining wall around the perimeter of the gap are extra cost items. In addition, special detailing for facial panel anchorage, stairs, elevator, and mechanical shafts are required at the isolator

level and are extra cost items.

In the soil system, the different .soil layer thicknesses and

stiffness and strength properties modify the dominant peri- ods of the earthquake ground motions input into the base of

the building. To be effective, the base isolation system must

shift (increase) the fundamental period of the fixed-base

structure to a period range where the eanhquake ground

motion input energy is relatively low. The soil subsystem

together with the characteristics of the incoming earthquake

ground motions control the period range in which the earth- quake ground shaking energy is dominant.

The potential benefits from base isolation of a building

include:

• Savings in the cost of the lateral force resisting system and nonstructural system due to reduction in the seismic design forces.

• Reduction of potential structural and nonstiuctural damage over the service life of the building. Minimizing damage to essential equipment, .services, and other

nonstructural elements.

"Some of the larger amplification factors can be 1.5 times or

larger than the average factors. 'To date, none of the known prototype ba.se isolated building

has been subjected to severe eanhquake ground shaking

near its upper design level.

415
• Reduction of potential down time required for repairs of earthquake damage.

• Ensures that the structure and its contents remain functional during and after a major eiirthquake.

The last two items are important in an economic analysis and can be quantified in dollars for civilian buildings where

losses from potential down time can be estimated from

available operating cost and revenue data. However, the U.S. Navy is a nonprofit organization and the coiresponding

operating cost and revenue data are not readily available. For this study, the last two items are considered as intangible

benefits from base isolation.

The purpose of this paper is to determine the economics of laminated rubber and steel bearing base isolation system for buildings from the very limited data available.

2. ESTIMATED COST OF LAMINATED RUBBER/

LEAD PLUG. ISOLATION BEARINGS

The cost of the base isolation bearings depends on the weight of the building excluding its foundation. This is because the isolation bearings must support the weight of the building under ordinary conditions. In addition, the beanngs

must be able to resist the earthqu;tke-induced inertia forces of the isolated building. These inertia forces increase with

the weight of the building supported by the isolation bearings. Hence, the cost of the base isolation bearings is ex- pected to be smaller for the lighter steel buildings as compared to the heavier reinforced concrete buildings.

In this study, estimated costs of the bearings expressed as a percentage of the unit cost (cost per square foot) of newly

constructed steel frame and concrete frame office buildings

in southern California.

The typical seismic design floor loads for steel frame and

concrete frame office buildings are as follows:

Steel Frame Concrete Frame

Finish l.Olb/ft^ Floor covering 1 Ib/ft" S tee I deck 3.1 Concrete frame 129 Concrete fill 32.0 Partitions 20

Steel bearing girders 5.9 Ceiling 5 Columns 1.5 Mech and Elect 5 Panitions 20.0 Exterior wall 4

Ceiling 10.0 Misc. Mech and Elect 5.0 168 Ib/ft^ Misc. _L0 Seismic L.L.=50/4=13

79.5 lb/ ft^ 181 lb/ft' Seismic L.L.=50/4= 12^

92.0 Ib/ft^

For practical purposes, the typical seismic design floor load

for concrete frame buildings is about twice that for steel frame buildings. For reference, the typical seismic design

floor load for small stucco and wooden frame apartment or

office buildings is about 44 lb/ft\ or about one half of that for steel frame buildings.

The estimated cost of the laminated rubber and steel/lead plug base isolation bearings for modem commercial build- ings is about $21/ton or $0.0105/lb of supported weight

(Ref 6). Multiplying the latter by the seismic design floor loads gives unit costs for ba.se isolation bearings at S0.97/ft^ and $1.90/ft^ respectively, for steel frame and concrete frame buildings. Ba.se isolation is more attractive for build- ings with smaller design fioor loads.

Based on the 1986 construction costs given in Reference 17, the estimated unit costs for different types of new construction in southern California are given in Table 1. Due to the difference in construction required to satisfy functional requirements, the unit costs for hospital, telephone ex- change, and medical clinics are much higher than the unit costs for steel frame and concrete frame office buildings.

Similarly, the unit costs for wood framed apartments or

office buildings, reinforced masonry or steel framed ware- houses or industrial buildings are less than those for steel framed or concrete framed office buildings.

The unit cost of the base isolation bearings for steel frame

and concrete frame buildings expressed as a percent of their corresponding unit construction cost are 1.1 % f $0.97/ft^ / $88/ft^) and 2.2% ($1.90/588), respectively.

For steel frame and concrete frame hospitals, these percent

costs for the base isolation bearings are 0.6% ($0.97/5166) and 1.1% ($1.90/5166), respectively. These values are about one-half of those values for steel frame and concrete frame office buildings. The percent costs of the base isola- tion bearings for the other buildings shown in Table 1 can be determined in a similar manner.

The percent cost of the base isolation bearings decreases

with increasing unit construction cost of the building. Thus,

base isolation tended to be economically more attractive for buildings with high unit construction costs than those with low ones.

3. ESTIMATED INCREMENTAL COST FOR INSTALLATION OF THE BASE ISOLATION SYSTEM

The incremental cost associated with the installation of a base isolation system in a building is the difference between

the added cost for the base isolation system and the savings

from the reduction in structural and nonstructural costs.

416
3.1 Added Cost % of Building Cost Cost

The added cost for the base isolation system includes the cost of the isolation bearings, basement slab above the base

isolators, retaining wall around the perimeter of the gap (see Figure 5), extra excavation for the gap, sliding cover for the gap, flexible utility connections, etc.

The William Qayton Building is a four-story (including

basement) reinforced concrete frame building of 97 meters

(318 feet) by 40 meters, (131 feet) in plan with column

spacing at 7.2 meters (24 feet), on centers constructed in New Zealand (Ref 4). It is the first building fitted with a laminated rubber and steel/lead plug base isolation system.

The building was designed for 1.5 times maximum ground

motion time-history from the 1940 EI Centro earthquake,

i.e., a peak horizontal acceleration of about 0.52g. The base

isolation system cost was about 4% of the total building

cost. With the base isolation system, the reinforcing detail- ing for the beam-to-column joints were relaxed somewhat.

Otherwise, beams 200 mm deeper would be necessary to accommodate the full joint shear reinforcement detailing

required. However, no definite indications were given in the reference on whether the building was more or less expensive than a conventional building. Hypothetically, if there were no savings in structural and nonstructural costs, the William Clayton Building would cost 4% more than a conventional building. The building would be economical

only if the base isolation system reduced the present worth

of expected future earthquake damage to the building by

equal to or greater than 4%. The cost of the base isolation bearings for concrete frame office buildings given earlier is 2.2% of the total building cost. Subtracting this from the added cost of the base isolation system of about 4% gives

1.8% as the approximate added cost of the base isolation system, excluding the cost of the bearings.

Kelly and Hodder (Ref 16) reported the estimated added

costs of the base isolation system and the savings in structural and nonstructural costs for a six-story 170,000 ft^, steel frame medical building. A figure showing the building and

the itemized costs from the reference is reproduced here as Figure 6. The cost of the 46 base isolators is given as $46, (XX). The author believes that a more probable cost should be about 170,000 ft^ x $0.97/ft^ = $165,000 for the isoladon bearings. In addition, a miscellaneous cost of $40,000 would be more appropriate than the $30,000 shown,

due to the increase in isolation bearing cost. The author's estimated total cost for the building is $14.96 x 10* (170,(X)0 ft^ x $88/ft^). The modified added costs for the base isolation system are:

46 Base Isolators $165,000 1.1%

Retaining Wall 144.000 0.9 Basement Slab Separation 57,(XW 0.4 Miscellaneous 40.(XX) 01

Total $406,000 2.7%

The added cost for the base isolation system excluding the cost of the base isolation bearings is 1.6%. This is approximately equal to the corresponding 1.8% determined earlier for the four-story concrete frame building. The slightly

higher percentage for the William Clayton Building might

be expected because it has only four stories as compared to the six stories for the building above. The fixed portion of

the added cost for the base isolation system excluding the bearings is distributed over a smaller number of stories and, hence, smaller total floor area. Thus, the resulting added

cost percentage is expected to be somewhat larger.

3.2 Savings in Structural and Nonstructural Cost

Becaase base isolation reduces the seismic forces on the building, the cost for providing seismic resistance to structural and nonstructural elements is expected to decrease as compared to a fixed-base building. However, the cost reduction is not expected to be great for two reasons.

First, the current practice is to design the base isolated building to remain essentially elastic for an equivalent base

shear* of between 0.20 and 0.25g associated with a maximum ground acceleration of about 0.5g. This is approximately equal to the code design base shear for seismic Zone 4 (e.g., UBC, SEAOC, NAVFAC P-355). The structural detailing for base isolated buildings can be relaxed somewhat. However, to fulfill the intent of base isolation - to minimize the building and its contents from damage due

to major earthquakes - the base isolated building is generally

designed to remain essentially elastic and with as much

reserve ductility and energy absorbing capacity as possible. This is done to provide an extra margin of safety in the rare event that the earthquake ground motions at the site are

larger than or much different from the characteristics of design ground motions. If the actual maximum eanhquake

ground acceleration is much larger than the design acceleration, the laminated rubber and steel/lead plug base isolation system would still remain functional, but the seismic forces transmitted to the building would be much increased (Fig- ure 7). The ductility and overstrength capacity (above the

elastic) of the lateral force resisting system of the building

will resist the increase in seismic forces by inelastic defor-

*Dynainic analysis procedures should be used for the design and analysis of base isolated buildings.

417
mation as in a conventional fixed-base building.

Second, the cost for providing seismic resistance to the

structural system of a multistory building is only a small

fraction of the total building cost. For illustration, a listing of the structural and nonstructural costs for a 13-story steel frame building (from Ref 18) is shown in Table 2. The total structural cost including seismic provisions for Zone 4 is about 30% of the total building cost. About one-fifth (5.7%)

of this 30% goes to providing seismic structural resistance. About 2.3% of the total building cost (near the bottom right comer of the table) goes to providing seismic resistance to nonstructural elements. Thus, the co.st of providing seismic

resistance to structural and nonstructural elements of a steel frame building is approximately equal to 8% of the total cost

for a conventional building. Any savings by base isolation would have to come from this 8%. Because of the desire for

essentially elastic response of the building, base isolation is expected to reduce this percentage by only a small amount. Most of the actual savings, however, would probably come

from the reduction in the sizes of the structural elements and

reduction in bracing and anchorage of the electrical and

mechanical components had the building been designed to remain essentially elastic under a maximum earthquake

design ground acceleration of between 0.4g and 0.5g without base isolation'. For reinforced concrete frame buildings,

the cost of providing seismic resistance to structural and

nonstructural elements in seismic Zone 4 is about 10%

(Ref 18).

It is hypothesized that to satisfy the essentially elastic crite- rion, the 8% and 10% seismic construction cost for steel frame and concrete frame buildings, respectively, could only

be reduced by about one-third as a result of base isolation. Hence, the savings in the structural and nonstructural cost

for steel frame and concrete frame buildings by using isola- tion are estimated to be about 2.7% and 3.3%, respectively.

The corresponding estimated reduction in structural and

nonstructural cost given for the six-story steel frame build- mg in Figure 6 is $384,000, or about 2.6% (5384,000/

SI 4.96 X 10*) of total building cost. This compares favorably with the 2.7% estimated previously. About six-tenths (1.5%) of this 2.6% cost savings is due to reduction in the

structural requirements that include the elimination of the shear walls required to reduce interstory drift and the reduction in requirements for floor and roof diaphragm connection requirements. The remaining 1.1% savings is due to the reduction in seismic bracing for mechanical and electrical components. Although a reinforced concrete frame building

would probably not require shear wall bracing to reduce

story drift or special floor and roof diaphragm connections,

savings in the structural cost would come from a reduction

in structural member sizes and steel reinforcement, and a

slight relaxation of the beam-to-column joint reinforcement

detailing requirement. It .seems reasonable to expect a

comparable savings in the structural cost for ba.se isolated concrete frame buildings of about 1.9% [(1.5/2.6) x (3.3%)].

The savings for the nonstructural cost of the concrete frame

building is about 1.4%.

3.3 Incremental Cost

For the four-story concrete frame William Qayton Building,

the estimated incremental cost is about 0.7%. That is. the estimated cost for the base isolated building is about 0.7%

larger than a conventional building without ba.se isolation. For the six-story steel frame building, the estimated incre- mental cost for the ba.se isolation system is about 0. 1 %. Due

to the approximate numbers u.sed, this incremental cost is rounded off to zero percent.

Base isolation for the four-story concrete frame William

Clayton Building would be economical only if base isolation could reduce the present worth of the expected earthquake , damage over the service life of the building by 0.7% or more.

4. REDUCTION IN EXPECTED FUTURE EARTHQUAKE DAMAGE OVER SERVICE LIFE OF THE

BUILDING

Reference 20 presented the present worth of expected future earthquake damage to buildings, based on average historic building damage versus earthquake intensity data (Ref 21)

using the probability versus maximum ground acceleration curve for an intermediate soil site in Califomia with a maximum ground acceleration of 0.4g, with 80% probability of not being exceeded in 50 years." The interest rate used was 10%.

The present worth of all the expected future earthquake

damage in 50 years for steel frame and concrete frame

buildings with seismic design is 2.7% and 4.2%, respectively, of the replacement cost of the building.

Assuming that base isolation is effective in reducing the expected earthquake damage by two-thirds, the present worth of the reduction in expected future earthquake damage

by base isolation for steel frame and concrete frame build- ings is 1.8% and 2.8%, respectively. Becau.se both of these values are larger than the corresponding incremental costs of 0% and 0.7% for the steel frame and concrete frame build- ings analyzed, respectively, base isolation is economical for both buildings.

'This is a requirement only for essential facilities that must

remain functional during and after a major earthquake. See NAVFACP-355.1 (Ref 19). 'This corresponds to a typical site in seismic Zone 4 (Ref

19).

418
A summary of the costs and benefits for the base isolated four-story concrete frame and six-story steel frame buildings

analyzed is shown in Table 3. The total benefits are 6. 1%

and 4.4%, respectively, for the concrete and steel buildings.

The larger benefit for the concrete building is due to its greater reduction in structural and .nonstructural costs for expected future earthquake damage by base isolation. The

total added costs for the base isolation system are 4.0% and

2.7%, respectively, for the concrete and steel buildings.

Apparently, the larger cost for the concrete building is due

primarily to its greater isolation bearing cost because of its greater dead weight. The benefits minus costs are 2. 1% and

1.7%, respectively, for the concrete and steel buildings. This suggests that base isolated concrete or steel frame

essential buildings are economically competitive if their cost

is not greater than about 2% of their conventional fixed base

counterparts.

The main benefits of earthquake base isolation for buildings

are the reduction of the expected future earthquake damage

to the building over its service life, ensuring that the build- ing and its contents remain functional during and after a major earthquake, and reducing the potential down time

required for repairs. Base isolation appears to be most

attractive economically and functionally for essential facilities, particularly those facilities that contain sensitive equip- ment whose cost is greater than the cost of the building. For

the latter buildings, the benefits from reducing future earth- quake damage to the building are magnified by the ratio of

building plus contents cost, divided by the cost of the building without contents, see Table 3.

5. CONCLUSIONS

Based on the results of the analysis presented, the following

conclusions are drawn:

1. Base isolation is economically more attractive for lighter

buildings with smaller design floor loads.

2. Base isolation is economically more attractive for build- ing structures with average or above average unit cost.

3. Base isolation is economical for low-rise steel frame and

concrete frame office buildings.

4. The main benefits of earthquake base isolation for a building are the reduction of the expected future earth- quake damage over its service life, ensuring that the building and its contents remain functional during and

after a major earthquake, and reducing the potential

down time required for repairs after a major earthquake.

5. Base isolation appears most attractive for essential

facilities such as hospitals, telephone exchanges, tele- communication centers, and computer centers, especially

those facilities that contain sensitive equipment that costs more than the facility itself.

6. RECOMMENDATION

It is recommended that earthquake base isolation be considered as an alternative to conventional fixed-base design and

con.struction for essential facilities located near potential

earthquake sources that must remain functional during and

after a major earthquake, particularly those that contain

sensitive equipment which costs more than the facility itself.

7. ACKNOWLEDGMENTS

The continual support by Mr. J.V. Tyrrell (Retired

NAVFAC Code 04BO), Mr. H. Nickerson (NAVFAC Code

04B2), and Mr. J. Cecilio (NAVFAC Chief Engineer) is appreciated.

This paper was sponsored by the Naval Facilities Engineering Command.

8. REFERENCES

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419
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16. Kelly, J.M. and Hodder, S.B. Earthquake Engineering Research Center. Report No. UCB/EERC-81/16:

"Experimental study of lead and elastomeric dampers

for base isolation systems." University of California at Berkeley, Berkeley, Calif., Oct 1981.

17. R.S. Means Co., Inc., "Building construction cost data 1986," 44th Edition, King.ston, Ma,ss., 1985.

18. Leslie, S.K. and Biggs, J.M. Massachusetts Institute of Technology. Report No. 341: Optimum seismic protection and building statistics, Mass., May 1972.

19. Departments of the Army, Navy, and Air Force. Army

Technical Manual No. 5-809-10-1, NAVFAC P-355.1, and Air Force Manual No. 88-3, Ch. 13.1, Final Manuscript' Seismic design guidelines for es.sential buildings (a supplement to Seismic Design for Buildings), prepared by URS/Jon A. Blume and Associates,

Engineers. Washington, DC, Mar 1985.

20. Lew, T.K. Naval Civil Engineering Laboratory. Technical Memorandum TM 51-86-10: Present worth

of expected earthquake damage based on historic and

rapid seismic analysis data. Port Hueneme, Calif., May 1986.

21. Sauter, F.F. "Damage prediction for earthquake

insurance," in Proceedings of the Second U.S. National Conference on Earthquake Engineering, Stanford University, Standford, Calif., 22-24 Aug 1979. Earth- quake Engineering Research Institute, Berkeley, Calif., 1979, p. 99-108.

13. Kitagawa, T.Y. and Hiro.sawa, M. "Base isolated building structures in Japan," proceedings of the 20th Meeting of the U.S. -Japan Cooperative Program in Natural Resources Panel on Wind and Seismic Effects,

N.J. Raufaste, Editor, NIST SP 760, National Institute of Standards and Technology, Gaithersburg, MD., Jan. 1989.

14. Lee, D.M. and Medland, I.e. "Estimation of base

isolated structure responses," Bulletin of New Zealand

National Society for Earthquake Engineering, vol II, no. 4, Dec 1978, p. 234-244.

15. Megget, L.M. "Analysis and design of ba.se isolated reinforced concrete frame building," Bulletin of New

Zealand Society for Earthquake Engineering, vol 11,

420
Table I. Typical J 986 Unit Costs for New Construction in Southern

California (from Ref 1 7)

Estimated Cost

Building Description ($/ft^)

Steel framed office building 88

Concrete framed office building 88

Hospital 166 Telephone Exchange 149

Police Station 128 Medical Qinics 94

Fire Station 88

Small stucco and wood frame 44-59

apartment or office building Reinforced concrete block warehouse 31-45

or industrial building

Steel framed warehouse or industrial 29

building

421
Table 2. Cost Increase for Increasing Seismic Force Design Levels

(after Ref 18)

Increase Over

Percentage Original Total Item Total Cost Construction Cost (%)

of Item for Zones

3 4

Structural

Structural Steel' 12.5 2.68 5.4" Foundation' 1.2 0.162 0.162 Concrete Walls

and Slab' 3.7 0.09 0.102 Composite Deck*

Total Structural 23.2 2.93 5.66" Nonstructural Masonry Core' 5.0 0.55 0.71

Precast Panels* 4.73 0.033 0.067 Plumbing 3.6 0.022 0.022 HVAC 18.8 0.15 0.197

Electrical (including lights) 8.5 0.104 0.104

Elevators 4.8 0.131 0.165 Window Systems 4.5 0.627'

Partitions 4.45 0.163 0.163

Acoustical Ceilings 2.2 0.114 0.163 Miscellaneous Metals il 0.089 0,089

Total Nonstructural 60.28 1.36 2.31 Total Code Design

(code design items) 32.93 3.51 6.44" Maximum Design*" 83.48 4.29 7.97"

*Code design.

"Using built-up members.

''New window system.

•^Total structural and nonstructural items.

422
Table 3. Summary of Costs and Benefits of Base Isolated Buildings Analyzed

Building Description Benefit (%)* Cost ($)*

Benefit- Cost (%.)*

\X/illi':jm (~^\o\jtr\r\ RtiilHinrt WlllldlH V_laYI-^ll DUIiUlIlg

four-story reinforced concrete frame

Reduction in structural and nonstructural costs3.3 Isolation bearings 2.2

Reduction in future earthquake damage2.8 6TBase isolation system

excluding bearings

1.8

4.0 2.1

Six-story steel frame

medical building

Reduction in structural and nonstructural costs2.6 Isolation bearings 1.1

Reduction in future earthquake damage

1.8 Base isolation system

excluding bearings

1.6

2.7 1.7

*Percent of total cost of building excluding contents.

423
>

PERIOD

SHIFT

/ *

\^

«m

D

a

a

K

il o

Period (sec)

Figure 1. Idealized base shear demand versus period curve.

424
Figure 2. Modal and combined response of a base isolated shear

beam type building.

425
Steel Cap

Plate X

V//////

V//////

I ///////

I///////

Cylindrical Lead Plug

Rubber

Steel Plate

Figure 3. Idealized cross section of laminated rubber and steel/lead

plug isolation bearing.

426
EL CENTRO SIGNAL

V» 0 H » 200

LEVEL

FLOORS

FLOOR 4

FLOOR 3

FLOOR 2

-

FLOOR I

BASE

TABLE

Fixed- Base — Isolated

I 2 3 HORIZONTAL ACCELERATION (g)

Figure 4. Floor accelerations for five-story, one-third scale model

steel frame building loaded by 1940 El Centro earthquake

time history with a scaled peak acceleration of 0.55g

(after Ref 16).

427
Sliding Grill

Retaining Wall

,StruCTural Frame

Foundation

Figure 5. Typical layout for base isolated building.

428
3

e

aC 0

.91 .81

1(1

429
LEAD FILLED BEARINGS

EL CENTRO SIGNAL V=0

.75

1.5-

TABLE ACCELERATION — FLOOR 5 ACCELERATION —• BASE ACCELERATION

200 300 400

HORIZONTAL SPAN

500

Figure 7. Nonlinear response with lead- fi 1 led bearings under

increasing earthquake intensity (from Ref 16).

430
Damages Caused by Storm Surges in Maldives

by

Takaaki UDA*

SUNftiARY Maldives islands in Indian Ocean were

inundated during April 10 and 14 in 1987 due to '

the slora surges. It is found by the analysis of -

the danage that land reclamation induces coastal

disasters in stora surges. !t is suggested that

the coastal developnent should be planned under

the deep consideration of the relation between

land use and storm surges.

KEY WORDS : Maldives. Coral reef, Stora surges.

Danages. Land reclaaation.

1. INTRODUCTION During April 10 through 14 in 1987 stora

surges attacked the Maldives islands located in

the Indian Ocean, and sone islands including the capital island (Male island) were inundated.

These islands are located in the tropical region. ..

and consist of around 1200 islands. Usually wide

coral reef develops around the islands and high

incident waves break on the edge of the wide reef.

For this reason it is considered that the coastal

disasters related to stora waves are less

frequent. Nevertheless large damages were caused

by the stora waves in Maldives. One of the

reasons of occurrence of this disaster is due to

the unreasonable land reclaaation along the shore

on the coral island. It is pointed out that the sufficient assessment is required before the

various coastal works.

2. COASTAL DISASTERS IN MALDIVES •

2.1 Location of Maldives islands

Maldives islands consist of about 1200 small

islands scattered between 0° 45' S and 7° N in about 73° E in the Indian Ocean as shown in

Fig. 1(a). The length of chain islands is 820km

long in the north and south directions and 130km

in the east and west directions. The shortest

distance to India is 480km, and the distance

between Male island and Colombo in Sri Lanka is

640kB. Main part of Maldives with many population

is shown in Fig. Kb). A group of atolls has a circular shape as a whole. North Male Atoll where

Maldives capital is located is shown in Fig. 1(c)

in Buch larger scale. Male island is a saall island located on the south end of North Male

Atoll and Hulule island with international airport

is next to this island. North Male Atoll itself

consists of aany saall atolls. The number of

islands appropriate for the habitation is around

200. 2.2 Land reclaaation plan in Male island

Total area of Maldives is 298kB^. and the population is 181,453 as of March in 1985. The capital is located on Male island, and aany people

(46,334 as of March in 1985) live in a saall

island. Originally the area of this island before

the land reclaaation was as saall as 108 ha, but

in resent years land reclaaation has been rapidly

carried out. Before 1979 land reclaaation of 14.9 ha was carried out(Fig. 2). Thick solid line in the

figure shows the original shoreline, and dotted

line is the revetaent line existed in 1979. Since

then population has continued to increase, and lack of the land proceeded. To this reason large

scale land reclaaation was started. Land

reclaaation was aainly conducted at the southern

part of the island, and reclaimed area extended up to the vicinity of the reef edge (Fig. 3). Reclaimed area attained 51.8 ha until November in

1985. This accounts for 87X of all the planned

land reclamation area (59.7 ha). The sediment amount necessary for the planned land reclamation

i 3 was about 85 10 m, and almost all of sediment was the dredged sand in the port located in the

northern part of Male island.

2. 3 Damages due to storm surges Storm surges began at 9 o'clock on April 10.

* Head. Coastal Eng. Div. , Public Works Res. Inst. Ministry of Construction.

431
On April 12 sea water began to retreat, but the sea level was still high enough. Finally

extraordinary high tide continued until the

morning of Apri 1 14. Meteorological observation at Hulule island

where the international airport is located showed

no strong wind during the disasters between April

9 through 13. The Meteorological Agency in Sri Lanka reported that overall tide condition did not change near Sri Lanka, and therefore storu surges

in Maldives would be local phenoaenon.

Inundated area in the southern part of Male

island is shown in Fig. 4. The inundated area extends all around the island, while centering the southern part of the island. Total inundated area attains 48X of all the area of this island. Since

this island is a coral island, the scan elevation

is as low as about 1. 5i above the lean sea level. Therefore wide area was inundated. It should be

noted that the inland boundary of the inundated area agrees well with the original shoreline

(Fig. 2) before the land reclaiation. It clearly

points out the weakness of the reclaaation area against stora surges. By this stora surges about

205; of sand (3.62 lO^a'') used in the reclaaation

was carried away. The daaage loss was about 1.

3

million dollars. Since all islands in Maldives are coaposed of coral islands, it is very

difficult to get sand for land reclaaation. and therefore the loss of sand becaae a large social

problea. Of all the five investigated

islands, Eaboodo and Finolu islands are in South

Male Atoll, and Haru and Little Haru Islands are

located on the outer ridge of North Male Atoll. By the comparison of daaages on each island,

damages on two islands located in the North Male

Atoll are smaller than those on the other islands.

On the other hand, large damages were observed at the chain of islands extending northeastward froa

Male island and forming the outer boundary of the

atoll as shown in Fig. 1(c). Furthermore, large

damages concentrated on the southern and

southeastern parts of the each island. Taking

these conditions into consideration, it is concluded that the storm waves attacked the

islands from southeast. The causes of this storm surges are considered

to be as follows. First, tidal range is 75cm in

spring tide and 25c» in neap tide at ColoBbo in Sri Lanka due to the tidal record measured between

1956 and 1962, and usually spring tide appears in

a period of March and April. Since the storm surge began from April 10, the tide in Maldives

was at spring tide condition. Secondly, incident wave height was at least larger than 2b, judging

the wave condition from the photographs or video

record by inspection. Then it is considered that the mean sea level on coral reef rose by wave setup and it accelerated the wave intrusion into the land. Although coral reef itself had a wave dissipation effect, coastal daaages were

accelerated because of the land reclaaation up to the vicinity of the edge of coral reef.

3. CONCLUDING REMARKS

This paper describes the coastal disasters due

to stora surges in Maldives. It was found through

the analysis of the disasters that the artificial

action (excess land reclaaation) strongly relates

to the natural disaster. Regarding the

development of coastal zone, it is required to have sufficient reserve against stora surges.

Froa this point it is concluded that not only the recovery of the daaages but also the consideration

of fundamental probleas regarding coastal zone aanageaent is iaportant for the prevention of the future coastal disasters.

REFERENCE

Uda, T. (1988) Field investigation on

damages caused by stora surge in Maldives.

35th Japanese Conf. on Coastal Eng.. pp. 212-

216 . (in Japanese)

(a) Location of Maldives

432
..L...:

^: K a n I f 1 n o 1 u

""^ B o n d o s Z^ ' Little H a r u

a n ^G' H s r u

' NORTH MALE

ATOLL

*;i k u r u m fa a / ^

JTh u 1 u I e Air Port

.-Emboodo

Ma I e

I 1 0 I u

1/ SOUTH MALE

ATOLL

(c) North and South Male Atoll

Fig. 1 Maldives islands.

Fig. 3 Reclamation on Male island

after 1979.

^ .

, . rcclilmcd ares

/////////////^ 1 llcr April. 1979

ymm^^^ x e c 1 a I me d area alter under planning

breakwater

Fig. 4 Inundated area on southern side

of Male island.

433
Performance of Structures in Hurricane Hugo

by

Richard D. Marshall*

ABSTRACT

Surface wind speeds and damage to structures

during t±ie passage of Hurricane Hugo throu^

the Virgin Islands and Puerto Rico are described. Because of the scarcity of actual

wind speed records and observations, it was necessary to resort to other indicators of

surface wind speed such as reconnaissance

aircraft data and intensity of damage. Maxinum

wind speeds on St. Croix corresponded to a mean

recurrence interval of approximately 300 years,

naking Hugo one of the more intense hurricanes

of the century. On St. Thoims and over northeastern Puerto Rico the wind speeds were

equal to or less than t±)e 50-year wind speed

for this region of the Caribbean. Damage in

the affected areas was extensive, in particular

those areas exposed to the northeastern

quadrant of tte stx)mi. The structural behavior

of certain higjirrise and single-story buildings

under overall wind loading was good, hbwever,

nnore attention needs to be given to the

attachment of nonstructural elements such as

doors, windows and cladding.

KEYWIRDS: Buildings; building codes; damage;

hurricane; tropical cyclone; wind; wind speed.

1. TNTRODUCnON

Hurricane Hugo passed throu^ the U.S. Virgin

Islands and over northeastern Puerto Rico on

Septenber 18, 1989. The probable storm track

is plotted in Figure 1 and the circles along

the track represent a typical eye diameter of

about 35 kn. Other important meteorological

characteristics of Hugo are presented at this

meeting in a companion paper by Golden (1990).

As usually happens following extrane events

such as Hugo, there were numerous reports of

extraordinarily hi^ wind speeds in the

affected areas. Problons encountered in

tracking some of the more spectacular claims to their source and the methods used to establish

probable maximun surface wind speeds from other

than standard measurements have been described

by Marshall (1990).

To assess the perfomance of buildings and

other structures, it is important that

representative surface wind speeds be

ascertained. In the case of Hurricane Hugo,

only tW3 verifiable wind speed records were obtained on Septenber 18, one at the Etoosevelt

Roads Naval Station directly southeast of

Ceiba, Puerto Rico, and the other at San Juan

International Airport. Other indirect sources of wind information used to estimate surface

wind speeds were fli^t data from hurricane

reconnaissance aircraft, observations of type

and intensity of damage, the response of

selected structures to wind effects, and

eyewitness accounts of the storm.

This paper presents a sumnary of the estinnted

and measured surface wind speeds as well as the

speeds corresponding to wind load design

requiranents contained in the relevant regional

building codes and standards. These speeds

provide the basis for assessments of structural

performance described in the remainder of the

paper.

2. SURFACE mm SPEEDS

Surface wind speeds in Hurricane Hugo are suimarized in Table 1 . Note that these speeds

represent SI conversions frcm the original

units of knots, possibly giving the impression

of a better resolution than is actually the case. These speeds have been adjusted for

standard conditions of a 10-meter heigjit in

flat, open terrain (airport exposure) using

techniques previously applied to wind records

obtained in Hurricane Alicia on the Texas Gulf

Coast (Marshall, 1984). It has been assumed

here that the ratio of peak gust speed to

* National Institute of Standards & Technology

Gaithersburg, Maryland 20899

434
sustained speed is approximately 1.25. This

ratio is consistent with wind speed records

obtained in Hurricane Hugo and in other

Atlantic and Gulf Coast hurricanes. For those

areas where the local terrain promotes

channeling or shielding, the actual speeds may

differ significantly fron those listed in Table

1. The code -specifled design speeds for the areas affected by Hugo are listed in Table 2,

along with the applicable code or standard and

year of adoption. For ease of ccnparison, the

specified design speeds have been converted to equivalent sustained speeds, i.e., wind speeds

averaged over 1 minute.

Wind speed statistics for the Caribbean have

been updated and refined by Davenport et al. (1985) and this source of information has been

used to characterize the severity of Hugo's

wind speeds in terms of their mean recurrence

intervals as listed in Table 1. For St. Croix

the probable maximin sustained speed of 57 iVs

corresponds to a mean recurrence interval of

about 300 years. For the island of Culebra

this interval is approximately 200 years for

the estimated speed of 54 n\/s. On Vieques the

probable maxinum sustained speed of 49 m/s

corresponds to an 85 year mean recurrence

interval. For St. Thomas and northeastern

Puerto Rico, the maxinum sustained speeds were

equal to or less than the 50 year speed of 45

m/s.

3. WIND Et^MAGE

Observed damage in the affected areas is in

general agreement with the surface wind sf)eeds

listed in Table 1. This damage ranged from

superficial to total devastation. In general,

the most damaging winds were located in the

northeast quadrant of tire storm. In the

following paragraphs, the damage is described

in sequential order, following the general path

taken by Hugo.

3.1 St. Croix

The strongest winds came from the northeast and

caused the heaviest damage along the north

coast from the Salt River eastward to the end

of the island. Damage in this area was

intensified by the local terrain viuch slopes

steeply upward to a central east-west ridge

running the length of the island. This same

ridge provided some shielding for stnjctxires

located on its south slope.

In the southwest sector of the islands near

Frederiksted the most damaging winds came from

the southwBst following passage of the eye. The airport and oil refinery are located in the

south central portion of the island vhere the

strongest winds were out of the northeast.

Hoiirever, the winds in this area were not as strong as those affecting the north coast near Christiansted where local terrain conditions

increased the surface wind speeds.

Figure 2 shews typical damage to a condominiun

located on the north coast, approximately 6 km

east of Christiansted. The building system

consists of concrete floor slabs and concrete

masonry walls with cast-in-place comer colunrans and perimeter beams. Balcony decks and roofs

are of wood construction. The strongest winds

from the northeast were approximately normal to

the long axis of the building.

Figure 3 shows a series of newly constructed

apartment buildings directly to the west of

Christiansted. Although the hip roofs and

their cladding suffered some damage along the

edges and ridges, the main problem appears to be inadequate roof- to-wall connections. In

fact, this mode of failure was conriDn

throu^iout St. Croix.

Figure 4 shews a failed metal building directly

southv^^t of Christiansted and approxinately

one forth of the way up the north slope of the

main east-wst ridge. Again, the terrain

undoubtedly caused an increase in wind speed

and the strongest winds were approximately

face-on to the failed wall. It appears that

the roof sheets and purlins v^re carried away

as a unit, leaving the end frame unsupported

and allowing the windward wall to collapse

irE«rard.

Figure 5 shews damage to petroleun storage

tanks at the oil refinery on the south coast of

St. Croix, approximately 3 km east of the

airport. Some reports put the wind speeds in excess of 95 td/s at this location, the hi^Test

of any surface speeds claimed for Hugo.

Howerver, a followup investigation concluded

that no wind speed measurements were obtained

here or at any other location on St. Croix.

435
Directly northfr^t of the airport is a piibiic

housing development with units of 2 or 3 stories in hei^t. Panelized concrete

construction is used throu^-iout and t±ie endwalis are fabricated of louvered inetal panels mounted in a perimeter channel. This

channel was attached to the concrete panels by

means of explosive -driven studs. As can be seen in Figure 6, the concrete panels suffered

no visible damage, but nuch of the louvered

endwall was removed by t±ie wind. Many of the

housing units in this development appear to have been unoccupied at the time of Hurricane

Hugo.

3.2 St. Thomas

Damage on St. Thomas was not as widespread as on St. Croix. However, the terrain is

substantially rou^ier than on St. Croix and

nuch of the stnnjctural damage appeared to be

the result of accelerating flows due to hills

and ridges. San Juan radar showed strong

convection bands over the island and the

possibility of locally int:ense winds due to microbursts (vertical momentum transport)

cannot be ruled out. The radar images

indicated that the strongest winds ranged from

northeast to southeast. Intensity of damage was similar to that observed in northeastern

Puerto Elico.

3.3 Culebra

In Puerto Rico, the island of Culebra

experienced the most severe damage, being only

sli^tly less intense than the damage observed on St. Croix. Figure 7 shews a housing area

adjacent to the local airport which was totally

destroyed. The quality of construction was poor and it is doubtful that there was any

attempt to conply with the territorial building

code. In addition, the t:erraln slopes steeply

upward to the west and this very likely caused an acceleration of the intense easterly winds

from Hugo. Directly to the south is the

Ensenada Honda where many small watercraft were

either sunk or were driven ashore by the strong

easterly winds.

3.4 Vieques

This island is to the south and west of the

storm track and, therefore, likely experienced

less severe winds than did Culebra. Figure 8 shows a housing area where the individual units were cleanly ranoved from their concrete postand-beara supporting systems. It was rot

possible to conduct a ground survey on Vieques

,

but it appears that the floor-to-frame

connection details in this case reed to be

inproved.

3.5 Puerto Rico

On Puerto Rico the sustained speeds ranged from

35 rVs at San Juan to 44 m/s at Roosevelt

Roads. Neither station experienced a clear

lull due to eye passage as did Cape San Juan at

t±ie extrone northeast tip of the island,

althou^ Roosevelt Roads came close.

Figure 9 shows two reinforced corcrete

buildings located on the east coast of Puerto

Rico between Cape San Juan and Fajardo. The

building on the ri^t was oriented so that the

strongest winds from the north were blowing

directly onto its broad face. Subsequent

inspection showed that many floors near the top

of this building had suffered failure of the

windward curtain walls and, subsequently, the

interior partition walls and leeward curtain

walls were completely blown out. Figure 10

shows an endwall of the building oriented so

that the strongest winds came from left to ri^t. Note the local damage to the perimeter

beams under the floor slabs caused by

transverse displacanent of the building. The

adjacent shearwalls showed no signs of

distress. By reference to ANSI A58.1 (1982),

the Puerto Rico Building Code has, until

recently, specified a reference wind speed of

41 m/s. In 1988 the code was revised and the

reference wind speed was increased to 50 m/s.

However, seismic rather than wind requirements

of the building code governed the structxiral

design of the buildings shewn in Figure 9.

The house shown in Figure 11 is typical of a

class of structure in Puerto Rico that

performed extranely veil in Hugo. The walls

are either cast-in-place concrete or are constructed of concrete masonry units with

integral reinforced concrete coluims and

perimeter beams. The roof slab is 100 to 125

rrni thick. As with the hi^irise concrete

buildings just described, the seismic

requirarients of the building code governed the

436
design and t±ie substantial dead loads make this

type of structure hi^ily resistant to wind

forces. Cbserved damage was limited to inadequate attaciinent of doors and louvered

window panels to the concrete walls and

perimeter beams.

A type of failure coranonly encountered in

northeastern Puerto Rico is shown in Figure 12. The corrugated steel sheet is attached to the

purlins by self-drilling/tapping screws. The

hi^-strength sheet is susceptible to low cycle

fatigue at the attachment points and develops

fatigue cracks after only a few minates of wind

action. This problem has been investigated and

reported by Morgan and Beck (1975). Use of a

large washer, shaped to confom to the ridge

contour of t±ke corrugated steel sheet has been

shown to be a sinple, cost-effecti-ve solution

to the problon.

4. CXMUJSIONS

For all of the regions affected by Hurricane

Hugo on Septa±)er 18, only twa verifiable wind

speed records ware obtained. Because of the

lack of data, it was necessary to rely on other

sources of information and aiiploy indirect

methods to obtain estimates of surface wind

speeds. Maxinun speeds on St. Croix correspond

to a mean recurrence interval of approximately

300 years. Estimated speeds on the islands of

Culebra and Vieques are consistent with mean

recurrence intervals of 200 and 85 years,

respectively. Speeds on St. Thomas and over northeastern Puerto Rico were either equal to or less than the 50 year wind. Widespread loss

of roof structures was observed on St. Croix

which suggests that code requirements for wind

uplift should be reviewed. In Puerto Rico,

both hi^nrise and single-story concrete

buildings that were designed to meet seismic

requironents per^onned well in Hugo. The

attachiEnt of nonstructural elanents such as

doors, windows and cladding needs tx) be

improved. Loss of corrugated steel sheet

roofing continues to be a widespread problen, even thou^ cost-effective solutions have been

available for several years.

5. ACKNOWLEDGMENTS

Photographs of damage were provided by Dr. Joseph H. Golden and Mr. John L. Vogel of the

National Weather Service. Dr. Mark D. Powell,

Atlantic Oceanographic and Meteorological

Laboratory, made available his analyses of

fligj-it data in the Caribbean. The author

collected information for this paper vMle a

me±ier of a post-disaster team organized by the

Comnittee on Natural Disasters, National

Research Council.

6. REFERHSICES

1. ANSI A58.1 (1982). "Mininun Design Loads

for Buildings and Other Structures,"

American National Standards Institute,

New York, NY.

2. Caribbean Uniform Building Code (1989). "Part 2 - Structural Design Requironents .

Caribbean Comnunity Secretariat,

Georgetown, Guyana.

3. Davenport, A.G. , et al. (1985). "A Hurricane Wind Risk Study for the Eastern

Caribbean, Jamaica and Belize With Special

Consideration to the Influence of

Topography." Report for the Pan-Caribbean

Disaster Prevention Preparedness Project.

4. Golden, J.H. (1990). "Meteorological Data

From Hurricane Hugo . " Paper presented at

22nd Joint Nfeeting, UJNR, Panel on Wind

and Seismic Effects, Gaithersburg, MD.

5. Marshall, R.D. (1990). "Hurricane Hugo in

the Caribbean - Assessment of Surface Wind

Speeds . " Conference : Six Months After

Hugo - Preliminary Findings . University of

Puerto Rico, Mayagpez, FR.

6. Marshall, R.D. (1984). "Fastest-Mile Wind

Speeds in Hurricane Alicia." NBS Technical

Note . No. 1197, NBS, Gaithersburg, MD.

7. Mor^, J.W. , and Beck, V.R. (1975).

"Sheet Metal Roof Failures by Repeated

Loading." Tech. Report No. 2, Housing

Research Branch, Australian Department of

Housing and Construction.

8. Puerto Rico Building Code (1968).

Regulations and Permits Adninistration,

San Juan, E^.

437
9. Virgin Is lards Building Code (1972).

Department of Planning and Natural

Resources, St. Thcmas, VI.

Table 1 . Probable Maxinun Wind Speeds in Hugp

LDC^CN SUSTAINED GUST MRI SPEED SPEED

(m/s) (w/s) (yrs)

St. Croix, VI 57 70 300

St. Ihccias 54 50

ViequBS, FR 49 59 85

Culebra 54 67 200

Roosevelt Roads 44 54* 50

San Juan 35* 41* <50

Table 2 . Specified Design Wind Speeds

HXmCN DESIGN SPEED CODE/STAND

Sustained (ra/s) (year)

U.S. Virgin Islands 40 VI BIdg. Code

(1972)

Puerto Rico 41 ANSI A58.1

(1982)

Pierto Rico 50 ER Bldg. Code

(1988)

VI/Hl (50-yr) 45 CUBC*

(1989)

" (100-yr) 50 tl

Figure 1. Hurricane Hugo Storm Track, September 18, 1990

438
u

439
Figure 4. Metal building, Christiansted , St. Croix

440
Figure 6. Concrete panel construction, St. Croix

Figure 7. Housing damage near Culebra airport

441
Figure 8 Housing damage on Vieques
Figure 10. Damaged edge beams, concrete highrise buildings

Faj ardo , Puerto Rico

Figure 11. Concrete housing unit, Puerto Rico

443
Figure 12. Failed corrugated steel sheet roofing, Puerto Rico

444
The Performance of Structures in Hurricane Hugo

1989 -The Carohnas

Peter ABSTRACT

A review is made of the performance of structures under the action of wind in

Hurricane Hugo which crossed the Carolinas on September 21-22, 1989. It is shown that in most locations wind speeds in the storm were

less than the design levels specified by local

building codes; but failure to incorporate

modern wind loading principles and to establish appropriate means of enforcing the

codes left South Carolina, in particular, illprepared for a storm of the strength of Hugo.

Extensive damage in such a storm was

inevitable and had been predicted.

Most damage was in the form of roof and wall

cladding failures which then led to extensive

rain damiage to the interior of the structure. Wood-frame construction performed well except

in exposed locations in the highest wind areas where loss of roofs, shear failures and

foundation failures were common. Unreinforced

masonry buildings performed very poorly, even

in sheltered locations. Older metal buildings

experienced considerable damage but newer ones built to increased wind-load requirements

performed better. The frames of steel and

concrete multi-story buildings performed well

but many suffered cladding damage even at relatively low wind speeds.

Recommendations are made to improve future

construction with regard to wind resistance,

including the establishment of a National Wind

Hazards Reduction Program by the United States Government

.

KEYWORDS: Wind; damage; structures;

hurricanes; building codes; wind speeds.

INTRODUCTION

Hugo was a fast moving hurricane which made

landfall on the mainland of the United States shortly after midnight on September 22, 1989. The center of the eye of the storm crossed the

coast at Sullivans Island, South Carolina, and

then moved across the state in a northwesterly direction. The storm weakened very

slowly. It was still inflicting considerable

damage when it entered North Carolina near the

city of Charlotte at 6:00 a.m. and eventually

caused tree damage as far north as the

Appalachian Mountains of Virginia (Figure 1).

Hugo naa the auDious aistinction or oeing tne

most expensive storm in U.S. history. Insured

losses due to wind damage are estimated to be

by

R. Sparks^

nearly S4 billion. To this must be added

insured flood losses of S500 million and

uninsured losses, including more than $1

billion in timber damage and over S300 million

in damage to military establishjnents

.

In most locations the wind speeds in Hugo were below the design levels specified in the

locally adopted building codes. This paper

examines the performance of structures under

the action of wind in the hurricane and, where

appropriate, seeks to explain their poor

performance

.

2. THE BUILDING CONTROL PROCESS

The United States does not have a national

building code. Responsibility for building

control lies with individual states and the

quality and type of control varies

considerably

.

In South Carolina the state exercises very

little control. Since 1972 local jurisdictions

have been permitted, but not required, to adopt the Standard Building Code. Once

adopted there is no requirement to update the

code nor any state guidance on its

enforcement. Any alterations to the code

however have to be approved by the state.

Some rural areas affected by Hugo had not adopted a building code.

The Standard Building Code (SBCCI 1974, 1982,

1988) is one of of three "model" building

codes used in the United States. It is

promulgated by the Southern Building Code

Congress, a democratic, non-government body

comprised mainly of building officials but few engineers

.

Although an American National Standard, (ANSI,

1972,1982) prepared by experts, exists for

wind loading, the Standard Building Code has

chosen only to use it after extensive

modification (Manning, 1985) . This has led to

a number of very serious shortcomings in the

code, primarily the non-inclusion of gust

effects, internal pressure, high suctions on the edges of roofs and near the corners of

Department of Civil Engineering, Clemson

University, Clemson, SC 29634-0911

445
buildings and increased wind velocities in areas adjacent to large bodies of water.

Some improvements have taken place in the

specification of wind loads in recent years

(SBCCI, 1988) but few buildings subjected to Hugo benefitted from these. Indeed many

buildings continue to be constructed without

reference to wind loads. Despite a long

history of poor performance in severe storms,

the code still permits certain forms of construction in wood and masonry which are inconsistent with its loading requirements.

Such prescriptive requirements are

particularly popular with architects who also

often select wall cladding and roof systems

without appropriate regard for wind loads.

A study of building practices in South

Carolina conducted in the mid-1980s (Sparks,

1988) warned that a hurricane, even below the

nominal design level, could create a multibillion dollar loss in the state. Hurricane

Hugo turned out to be that storm.

North Carolina has a tighter form of building

control. There is a state building code

(NCBCC, 1982), building inspectors and

officials must be certified and all plans for

major buildings are checked at the state

level . However, in common with the Standard

Building Code, the North Carolina State

Building Code underestimates cladding loads near the edges of roofs and corners of buildings

.

3. COMPARISON OF WIND CONDITIONS IN HUGO WITH DESIGN CQNPITIQNS

Building codes and standards in the United

States use the fastest-mile wind-speed,

measured at 33 ft. in open country, as their

reference. These speeds are based on the time

taken for a "mile" of wind to pass an observation point as recorded on a now obsolete instrument. In a hurricane such an instrument would average the wind speed over periods ranging from about 40 seconds to 1 minute

.

None of the anemometers which recorded wind

speeds in Hurricane Hugo measured a fastestmile and only one made measurements close to

33 ft. The fastest-mile wind speeds reported

in Table 1 are based on standard corrections

for averaging time, height of anemometer and

approach terrain, (Simiu and Scanlan, 1986,

Cook, 1985) These corrections can introduce

significant errors. Where several anemometers existed in one city, as in Charleston, greater

weighting was given to airport locations with

continuous anemometer traces. Measurements

were made at or near each of the locations

listed except as indicated in the table. The

speeds given at locations where no measurements were made are based on correlations with aircraft measurements and

comparisons with nearby ground stations.

The recurrence intervals listed are based on data assembled from a number of analyses of measured wind speeds and computer simulations

carried out in the last 20 years. These

values could be subject to considerable error. They should only be used to judge the relative

severity of the storm.

The design wind speeds given in Table 1 are taken from the Standard Building Code (1973-

1986) . They are fastest-mile wind-speeds with

a 100 year recurrence interval. However the

code used these speeds as gust wind speeds and

so, in effect, was using fastest-mile wind

speeds approximately 25% lower or a mean

recurrence interval of about 25 years. The

code also made no allowance for the higher

wind speeds experienced by areas adjacent to large bodies of water. With the rough seas of

a hurricane these would be about 16% higher

than those in open country. The design

conditions for onshore winds therefore

represented a recurrence interval of less than

15 years. The code also made no allowance for

the reduced wind speeds in suburban areas. In those areas the design wind conditions

probably did represent a recurrence interval

approaching 100 years.

These recurrence intervals tend to represent

upper bounds with regard to loads because the

pressure coefficients for wall and roof

cladding, used by the Standard Building Code

to convert wind speeds to wind pressures, were often less than half those used by the

American National Standard and many foreign

standards.

It is apparent that while Hurricane Hugo was nominally less than the storm for which

jurisdictions thought they were preparing

through the use of the Standard Building Code,

in actual fact the conditions, particularly in exposed areas, considerably exceeded the

design level

.

North Carolina uses a gust wind speed of 80 mph for all areas affected by Hugo. This was exceeded by about 20% in the Charlotte area

and wind conditions equal to the design level

may have been experienced as far north as the

Virginia border.

4 . SURVEY OF WIND DAMAGE

Wind damage to buildings and other structures

was observed along the South Carolina coast

from Edisto Beach to the North Carolina border

and inland to beyond Charlotte. Damage to roof

coverings, wall cladding and signs occurred

446
where the fastest -mile wind speed exceeded 60 mph . Major structural damage, including loss

of roof structure, collapse of single story

masonry buildings, complete destruction of mobile homes and extensive damage to woodframed construction and older pre-engineered

metal buildings, took place where the fastestmile wind speed exceeded 85 mph. However,

within these areas, damage was dependent on the upwind terrain. Damage occurred first in

exposed locations adjacent to the sea and in open terrain inland. In sheltered locations

even old wood-frame houses remained completely

undamaged where the fastest-mile wind speed

exceeded 3 00 mph.

North and South Carolina are heavily forested states. The trees provided significant

protection from the wind and undoubtedly

reduced the amount of damage to buildings.

However pine trees apparently began to fall as the fastest-mile wind-speed approached 60 mph

and many buildings, although protected from

the wind by the trees, subsequently suffered

damage from them.

In the Bull Bay area, where few trees were

left standing, the exposure of buildings would

have changed significantly during the course

of the storm.

Two other factors greatly reduced the damage.

Firstly, the densely populated islands of

Folly Beach, Sullivans Island and Isle of Palms experienced the eye of the storm. This meant that the strongest winds would have come from the north-east and south-west i.e. along

the shore. Many ocean-front buildings would

therefore have received shelter from

neighboring buildings. Secondly, the

strongest winds in the storm affected the most lightly populated area of the coast, occupied

primarily by a wildlife refuge. In this area

all settlement is landward of the Intracoastal

Waterway in forested areas. Had the storm hit

20 miles to the south, Charleston and the

adjacent islands would have experienced

onshore winds producing wind loads on oceanfront properties perhaps double those actually

experienced. Had the storm hit 50 miles to the north, all of the buildings on the densely

populated northern coast would have

experienced wind loads more than twice those

actually experienced.

4 . 1 One and Two Family Dwellings

In general these performed extremely well,

probably more as a result of shelter than any other factor. Major structural damage, nearly

always in exposed locations, resulted from

loss of roofs with subsequent collapse of walls, shear failures and foundation failures.

Roof failures were far less common than in recent hurricanes, probably due to the use of steep-pitched roofs and hurricane anchors.

However, in the highest wind areas, failures

occurred even where hurricane anchors had been

used, particularly after window damage

(Figure 2) . The improved roof performance

enabled the houses to survive long enough to

show up other weaknesses.

Failure to provide proper bracing led to some

shear failures in two story buildings

(Figure 3) . The traditional bracing techniques

proved to be insufficient, especially in

buildings with few interior cross-walls and

large openings in exterior walls.

Foundation failures due to wind were much more common in this hurricane than in other recent

ones. In most coastal flood-prone areas

buildings are elevated on wood or reinforced

concrete piles. On the South Carolina coast many buildings are constructed on concrete

block piers. The Standard Building Code

permits unreinforced masonry piers to have a height to width ratio of 10, if filled with concrete. Piers meeting this requirement

failed, as did those that were lightly

reinforced (Figure 4) . In meeting the needs

of flood insurance requirements, the very

large forces applied to the piers by the wind

acting on the building had apparently been

overlooked.

Many one and two-family dwellings lost

shingles. These apparently minor wind

failures resulted in significant rain damage

often magnifying the initial damage by a factor of 10. Repeated on an enormous scale,

this contributed significantly to the total

cost of the storm.

In a few areas wind and surge damage to single-family dwellings was difficult to differentiate, particularly in older buildings

built on grade or newer elevated buildings

subjected to surge in excess of the design

level, as in the northern part of the Isle of Palms and in Bull Bay. In some instances wind

debris was transported by the surge, making

the nature of the wind failure difficult to determine

.

Based on eye-witness reports and computer

simulations, it appears that on Folly Beach,

Sullivans Island and Isle of Palms, the strong

winds from the north-easterly direction before

the eye would have taken place while the

islands were dry. The surge rose rapidly

during the eye and would have been falling as winds rose from the south-westerly direction.

Several types of failure were possible and

observed

.

1. Due entirely to north-easterly

winds

.

2. Initiated by north-easterly winds

augmented by later surge action.

3. Initiated by surge action,

augmented by later south-westerly

winds.

In the Bull Bay area strong winds, initially

from the north-east, would have veered to the

south-east before the surge reached any

elevated houses. Most surge action would have

taken place as the winds veered to the south

and began to diminish. Since the surge was several feet above the elevation level

required to obtain flood insurance, some

houses adjacent to the Intracoastal Waterway

would have been subject to wave action while

others farther inland would have been

subjected to considerable upthrust . What

447
condition these houses were in as the result

of wind action prior to surge action is difficult to determine.

From Pawleys Island north almost ail major

structural damage to single -f ami ly dwellings

was due to surge action.

4 . 2 Multi-Family Wood-Framed Construction

Several of these professionally designed

buildings in exposed locations suffered

extensive roof failures, often precipitated by

window damage, (Figures 5 and 5) . In some of these buildings hurricane anchors had been

used to secure the roof. Whether or not these

had been selected after a structural analysis

is not known, but consideration of internal

pressure would not have been required by the

building code and the uplift loads could have

exceeded the design value after the failure of windward windows.

4 . 3 Masonry-Walled Buildings

Failure of masonry walls was observed inland

as far as Charlotte and complete structural

collapse was observed as far inland as Sumter,

(Figures 7 and 8) . All collapses took place in buildings using unreinforced concrete block or brick walls which would have met the height to thickness requirements of the Standard

Building Code. In most cases the roof to wall

anchors would also have met the requirements

of that document. Failures of this type of

structure have occurred in several recent

hurricanes (Savage et al . , 1984 , Mitchell et

al. 1986, Sparks et al., 1990). It was interesting to note that in this hurricane

most of the failures were not in buildings of recent construction. Discussions with

building officials in the Charleston area revealed that they no longer permit

unreinforced masonry walls in their

jurisdictions

.

A

.

4 Pre-Engineered Metal Buildings

A number of these buildings located in exposed areas were seriously damaged, (Figure 9) . These structures are very sensitive to the

deficiencies of the older versions of the

Standard Building Code. Buildings designed to the new wind-loading provisions appeared to have performed much better.

4 . 5 Steel and Concrete Framed Buildings

As observed in other hurricanes, the frames of buildings designed by structural engineers to meet the wind-load requirements in the

building code performed well. The wall

cladding, roof system and roof covering, often

selected by architects without reference to the wind-load requirements, performed very

poorly

.

An exterior wall system employed extensively

in the area uses gypsum board attached to metal studs. The gypsum board is covered with

insulation and stucco. Based on published

data on the ability of such systems to resist

wind suction, these systems should not have

been used on anv buildina over 30 ft. on the

South Carolina coast, even if designed in

accordance with the older versions of the. Standard Building Code. That code actually

underestimated the loads on the corners of

buildings by factors of 2 to 3. Nevertheless

this system was used on hospitals, the Red

Cross Headquarters, educational facilities,

condominiums, office buildings and hotels in the Charleston area and almost all high-rise

buildings in the Myrtle Beach area. Failures

were found on the coast as far north as the

North Carolina border where the actual wind

speeds were less than half of the design

value, (Figure 10) . Several buildings in the

Charleston area, including The Trident Medical

Center and the Red Cross Headquarters, were

seriously damaged. Window systems also failed

due to both positive pressure and suction.

Failures of flat-roof coverings, both buildup-roofs and single-ply systems, sometimes

coupled with the roof-deck failures, were also

to be found along the South Carolina coast as far as the North Carolina border. They were found inland as far as Charlotte. These

failures, which included schools used as emergency shelters, hospitals, and emergency

operations centers, resulted in extensive rain

damage to the interior structure and contents

of buildings. Based on a limited survey of insurance data, the rain damage in high-rise

buildings was often 20 to 30 times the actual

wind damage to the structure and multi-million

dollar claims occurred in areas where the storm had a recurrence interval of less than

10 years.

The Standard Building Code in use when most of these buildings were designed seriously

underestimated the uplift loads on the edges

of roofs. The design loads were almost

certainly exceeded in many places in Hugo. In some cases the shortcomings in the building

code may have led to the selection of an inadequate system, but in many cases the

system was probably chosen on the basis of cost rather than wind-uplift resistance.

4 . 6 Mobile Homes

A significant proportion of the population of South Carolina lives in mobile homes. The

failure rate of these homes was much greater

than conventionally built homes and the number

of deaths and injuries in the storm would

probably have been much higher had the

Governor not ordered the evacuation of all

mobile-home dwellers in the coastal counties,

(Figure 11 )

.

Many mobile-home owners were either uninsured

or underinsured and did not have the financial resources to face a storm like Hugo. It was this section of the community which was probably the most seriously affected by the

storm.

As mobile-home communities rebuilt they were forced to conform to many new government

regulations better suited to more permanent

dwellings. For example, in order to obtain

flood insurance mobile homes must be elevated

above the base flood elevation. After Hugo

this was being done by placing them on concrete block piers, in some cases more than

448
8 ft tali, making them very susceptibie to wind damage, (Figure 12).

5. rnNCLUSinN.S and RF.romENDATIONS

Hugo did more damage than other hurricanes,

not because of its severity as it crossed the

coast, but because it -affected a major

metroDolitan area and retained much of its

strength as it moved inland. However, it should also be noted that the worst conditions

were experienced by a very sparsely populated

area and the damage could have been very much

worse

.

In most countries national governments accept

the responsibility for developing and

maintaining building codes. With regard to wind effects on buildings, these codes usually

incorporate or refer to national loading

standards prepared by experts but subject to public review.

In the United States, where hurricanes and

tornadoes create one of the most hazardous

wind climates in the world, the Federal

Government has passed the responsibility for

building codes to the states. Many of the states have passed that responsibility on to local jurisdictions. Many of these

jurisdictions use model building codes which,

until recently, have chosen to ignore or significantly modify the American National

Standard A58.1, a standard prepared on the same lines as other national loading

standards

.

Although the model building codes are democratic non-government bodies, they have

been encouraged by the Federal Government to set performance standards in the belief that

the older prescriptive codes were too

restrictive. Setting performance standards

works well when a building is subjected to a structural analysis, but the vast majority of buildings- damaged by wind have never been

subjected to a complete structural analysis.

In such cases performance-based codes are unenforceable, unless proper deemed-to-comply

standards are available. Although such

standards have been available in North

Carolina and South Florida for many years, they are not yet available in areas such as South Carolina which are served by the

Standard Building Code. The prescriptive

construction requirements which do appear in that code are often inconsistent with the

loading requirements. Their use has led to many buildings being very vulnerable to wind

damage . ...

Much of the wind damage to buildings in

Hurricane Hugo was avoidable. The state of knowledge of wind effects was such that no building constructed after the mid-1960s need

have been damaged by the storm; but, for reasons more political than technical, South

Carolina found itself with a large population

of buildings vulnerable to wind damage.

Owners, insurers and government agencies must expect this situation to develop in other

states which have adopted similar building

control practices. In the future extensive

damage must be anticipated even in storms

below the desion level.

To reduce wind damage and loss of life in

future storms the following recommendations

are made.

1. The United States Government

should take an active role in reducing the wind hazard through a National Wind Hazard Reduction

Program. The Government has been

very successful in reducing the risks associated with flood and

earthquakes but has paid little

attention to the wind hazard.

Direct involvement in a national

code would be desirable but may

be politically unacceptable.

2. In the absence of a national

building code, states should

mandate the use of model building

codes and set up properly financed

code enforcement procedures.

3. Model building codes should adopt

the American National Standard

A58 . 1 and update their codes

as new editions of the standard

are issued.

4 . Model building codes should

introduce deemed-to-comply

standards for non-engineered

structures which are consistent

with their wind loading

requirements. Existing

inconsistent prescriptive

requirements for wood and masonry construction should be removed.

5. Insurers should recognize that in the past they have put unjustified

faith in the building control

process. As a result they are now insuring buildings which have a very high probability of wind

damage and consequent rain damage.

However, insurers should also note

that not all structural systems

are equally vulnerable to damage. Insurance premiums should reflect

this. Premium rates should also

be used to encourage good new construction .

6. Public officials should recognize

that some buildings which must continue to function during

severe storms, including

hospitals, emergency operations

centers, and public shelters are very vulnerable to wind damage.

All such buildings should be checked to ensure that they are capable of carrying the wind

loads specified in ANSI A58.1 and

strengthened if necessary.

7. Public officials should draw up plans to evacuate people from

certain classes of structure in

hurricanes, even if there is no threat of flooding. In the past.

449
evacuation has been ordered from

areas ptone to flooding. In Hugo

people were told to evacuate from

mobile homes whether or not they

were subject to flooding. This

probably saved many lives. In future storms evacuation should be ordered from certain other classes

of building which are very vulnerable to wind damage.

8. The Federal Flood Insurance Program should recognize that

flooding and high winds are not independent events in coastal

regions. Regulations to reduce

one hazard should not increase the other. Raising mobile homes 8 to

10 ft on masonry piers transforms

a relatively low flood risk, into a very high and dangerous wind risk. To avoid this sort of problem,

and difficulties in settling

insurance claims where buildings

are subjected to wind and flood

damage, comprehensive storm

insurance should be developed,

backed, if necessary, by the

Federal Government but contingent

upon good construction standards.

9. Educational and advisory

programs should be developed so that the construction industry can benefit from, wind engineering

knowledge. These programs should

include owners, insurers, and

mortgagors so that they can understand the risks associated

with certain forms of construction

and make informed financial

decisions regarding those risks.

10. Wind engineering research is many

years ahead of wind engineering

practice in many areas, but research is urgently needed on sheltering effects and the risks

associated with the selection of

particular structural systems.

The results of this research will

enable the consequences of past

construction practices to be

dete rmined

.

Finally it is worth recalling that the purpose

of building code enforcement is the protection

of the health, safety and welfare of the

population. Of the many thousands of

buildings affected by Hurricane Hugo, probably

less than li collapsed due to wind effects in

a manner that posed any threat to the health

and safety of the population. Most of these

were residences in evacuated areas, commercial structures and mobile homes. Thus by good

planning and some good design, Hugo posed a very small threat to the health and safety of the population. However, due to a lack of planning and poor design, it posed a large

threat to the welfare of the population.

Utilities were interrupted, in some places for

several weeks, and many people were left

unemployed because their places of work were

unusable. Many homes which suffered

relatively minor damage to their roofs an4

windows were uninhabitable for many months.

Those responsible for building control should

take note of this and enforce appropriate

regulations so that the welfare of the

population is protected and the community can recover rapidly after such a storm.

6. ACKNOWLEDGEMENTS

The author would like to thank Mark Powell,

NOAA, and Richard Marshall, NIST, for their

contributions to the establishment of the wind

conditions in Hurricane Hugo.

7 . REFERENCES

1. ANSI, 1972, 1982, "Minimum Design Loads

for Buildings and Other Structures,"

ANSI-A58.1, American National Standards

Institute, New York, New York.

2. Manning, B.R., 1985, "Hurricane

Requirements of the Standard Building

Code," in "Hurricane Alicia: One Year

Later, " American Society of Civil

Engineers, New York, New York.

3. Cook, N.J., 1985, "The Designers Guide

to Wind Loading of Building Structures,

Butterworths , London, U.K.

4. Mitchell, J.K., Abdel-Gaf f ar, A.M.,

Gentry, R.C., Leatherman, S.P., Sparks,

P.R., 1986, "Hurricane Diana, North

Carolina, September 10-14," National

Academy Press, Washington, D.C.

5. NCBCC, 1982, "The North Carolina State Building Code, " North Carolina Building

Code Council, Raleigh, North Carolina.

6. Savage, P.R., Baker, J., Golden, J.H.,

Kareem, A., Manning, B.R., 1984,

"Hurricane Alicia, Galveston and

Houston, Texas, August 17-18, 1983,"

National Academy Press, Washington, D.C.

7. SBCCI, 1974, 1982, 1988, "The Standard

Building Code, " Southern Building Code

Congress International, Birmingham,

Alabama

.

8. Simiu, E. and Scanlan, R.H., 1986, "Wind

Effects on Structures," Wiley, New York, New York.

9. Sparks, P.R., 1988, "The Risk of Hurricane Wind Damage on the South

Carolina Coast," a white paper issued by

the Coastal Hazards Advisory and

Mitigation Project, Clemson University,

Clemson, South Carolina.

10. Sparks, P.R., Baker, J., Bellville; J.D. and Perry, D.C, 1990, "Hurricane Elena,

Gulf Coast, August 29 - September 2, 1985," National Academy Press,

Washington, D.C.

450
I

I Figure 1. Track of Hurricane Hugo.

451
452
Figure 4. Masonry Pier Failure, Isle of Palms.

453
igure 7. Collapse of Masonry Commercial Structure,

North Charleston

454
Figure 8. Collapse of Masonry Commercial Structure, Sumter.

Figure 9. Damage to Metal Building, Charleston.'

455
456
457
I
Summary of Task Committee

Workshop Reports

Summary of Task Committee Workshop Reports

TASK COMMITTEES "C" AND "D"

REPORT OF U.S. -JAPAN ItlORKSHOP

REPAIR, RETROFIT, AND EVALUATION OF STRUCTURES

liJIND AND EARTHQUAKE RESISTANCE

MAY 12 - 14, 1930

A. Activities of the Uorkshop:

1 . Three sessions were held and 17 papers were presented

on the folloujing topics:

Evaluation of Structures.

1-1 Evaluation Procedures For Existing Buildings.

1-2 Aseismic Countermeasures For Existing Buildings.

1-3 Earthquake Hazard in Regions of Moderate

Seismicity

.

1-4 Comparison of U.S. and Japan Aseismic Designs.

PerFormance oF Existing Structures.

2-1 PerFormance oF strengthened R/C Buildings.

2-2 PerFormance oF Buildings in Silicon Valley.

2-3 Seismic Evaluation oF Existing Buildings.

2-4 Implications oF Loma Prieta on Building Codes.

2-5 Inspection, Evaluation, Restoration oF Earthquake

Damaged Buildings.

Research on Techniques oF Repair and RetroFit.

3-1 Research on Repair S Strengthening oF R/C Struct.

3-2 Guide For Repair and RetroFit Design oF R/C

Bldgs

.

3-3 Research on Repair & RetroFit oF Steel Structures

3-4 R/C Buildings Strengthened with Steel riembers

.

3-5 Shear Strength oF Anchor Bolts.

3-6 Seismic Strengthening oF R/C OFFice Buildings.

3-7 Recent Research on Strengthening by Steel Bracing

3-B Mortar Joints Between R/C Frame and Steel Bracing

2. Another presentation was added on the topic oF "NSF

Research on Quantitative Non-Destructive Evaluation

CQNDE) and Wind Engineering in the Year 2000.

3. Proceedings will be assembled and edited by J. Jirsa

oF the University oF Texas, and will be submitted to

the National Institute oF Standards and Technology

CNIST) For publishing. Included in the proceedings

will be discussions, questions, and any modlFicat ions

by the reporters or Task Committee Chairmen.

461
4, The follouiing documents were exchanged during the

workshop, or will be furnished to participants soon

after the Joint U.S.- Japan Panel Meeting:

a. Applied Technology Council CATC) Reports No. 21 &

SS on "Evaluation Procedures for Existing

Buildings . "

b. . NIST Special Publication No. 778 CJan 1990) on

"Performance of Structures During the Loma Prieta

Earthquake of October 17,1909."

c. Japan TCCMAR "Design Guidelines for Medium-Rise

Reinforced Masonry Buildings," which will be

finalized after July 1990. The U.S. Side will

attempt to have it translated into English.

d. Mailing List of Japanese "International Institute

of Seismology and Earthquake Engineering," so that

copies of workshop proceedings can be furnished to

each member.

5. It was recommended by workshop participants that future

workshops be held in conjunction with U.S. - Japan

Joint Panel Meetings on Ulind and Seismic Effects. The

Japanese Side recommended that the workshop proposed

for 1991 in Tsukuba be more focused on a specific

theme, to be determined by the Japanese members. Then a specific "Uind Engineering - Structural Performance

Evaluation Workshop" will be planned to be held in the

U.S. in 1992.

462
TASK COMMITTEE "J

REPORT ON THE SIXTH U.S. -JAPAN BRIDGE ENGINEERING WORKSHOP

ANALYSIS, DESIGN, PERFORMANCE AND STRENGTHENING

LAKE TAHOE, NEVADA, U.S.A.

May 7 and 8, 1990 c;.

The following resolutions are hereby adopted:

1. The sixth U.S. -Japan Bridge Engineering Workshop provided valuable

exchange of technical information which was beneficial to the bridge

engineering communities of both countries. Forty seven participants

representing government, academia, industry and private practice presented

41 papers in eight technical sessions. Main themes included: 1) "Wind

Effects and Wind Resistant Design;" 2) "Retrofitting and Seismic Design;"

3) "Cable-Supported Bridges: Construction, Performance and

Strengthening;" 4) "Seismic Performance and Analysis of Damaged Bridges;"

5) "Bridge Inspection, Repair, Testing, and Evaluation," 6) "Experimental

Investigations and Computer Modelling;" 7) "Structural Monitoring,

Diagnosis, System Identification and Non-Destructive Techniques;" and

8) "New Concepts, Dynamic Control, Seismic Isolation and Innovations in

Construction." In view of the importance of cooperative programs on the

subject of wind and seismic effects on bridges and on performance,

strengthening, structural monitoring, safety, nondestructive evaluation

and design innovation of these transportation structures, the continuation

of technical cooperation and interaction is considered essential.

2. The Workshop recognizes the importance of continued exchange of personnel,

technical information, research data and bridge response data from

significant earthquakes and strong winds, and use of available unique

research facilities in both countries. ; , ,

3. The primary purpose of the Workshop is to exchange technical information

on a timely basis that is usable by the bridge engineering profession.

The specific objectives of this Workshop are:

3.1 To ensure a first-hand exchange of information resulting from the

unique problems each country is faced with; for example, the

preliminary studies of the structural performance during the

October 17, 1989, Loma Prieta earthquake and the recent full-scale

vibration studies of the Yokohama Bay Cable-Stayed Bridge.

3.2 To stimulate more collaborative research work in this important area

of bridge engineering for better performance, rehabilitations and

strengthening of bridge structures in both countries.

3.3 To facilitate the exchange of technical information and ideas and to

promote the broadening of viewpoints for those involved.

4. The Workshop identifies the following technical topics for continued

cooperative research programs:

4.1 Dynamic control and seismic isolation of bridge structures including:

active and passive control, damping augmentation, experimental

studies and application to long-span cable-supported bridges.

463
4.2 System identification techniques for wind and seismic engineering

including: data interpretation studies, development of mathematical

models, identification of non-linear systems, reduced-order models,

computational algorithms and optimum sensor locations,

4.3 Non-destructive evaluation techniques including damage detection by sensor monitoring, expert systems, inspection techniques, structural

monitoring, condition assessment for safety, performance evaluation

and repair procedures,

4.4 Experimental and analytical studies on the effects of dynamic soil

characteristics, including bridge-soil interactions, on the behavior

of bridge structures.

5. The Workshop identifies the following technical topics for future

cooperative technical programs:

5.1 Retrofitting and strengthening methodologies for bridge structures.

5.2 Development of design criteria based on limit-state concepts.

6. It is anticipated that the interaction among the delegates from the United

States and Japan during the round-table discussions and the technical

study tour will lead to an enhanced exchange of data and innovative ideas,

and will also lead to the establishment of cooperative research efforts,

both theoretical and experimental, on bridge engineering. Because of the

importance, benefits and potential of these Workshops, it is recommended

to conduct the 7th bridge engineering workshop, just prior to the 23rd

UJNR joint meeting in Tsukuba, Japan, in May 1991. A study-tour will be

planned to visit research laboratories and bridge-construction sites

including the Honshu-Shi koku bridge site, and the base-isolated bridge.

464
Appendix:

Task Committees

A-J Reports

Appendix: Task Committee A-J Reports

Report of Task Committee on

(A) STRONG-MOTION INSTRUMENT ARRAYS AND DATA

Date ; May 15, 1990

Place : National Institute of Standards and Technology

Gaithersburg, Maryland 2 0899

Attendees: Japan Side T. Uwabe (Acting Chairman)

K. Ohtani

K. Tamura

(PHRI)

(NRCDP)

(PWRI)

U.S. Side A. G. Brady (Chairman)

A. G. Franklin

H. Meyers

(USGS)

(USAEWES)

(NOAA)

MISSION STATEMENT

1. Objective : The objective of the Task Committee is to coordinate the provision of

strong-motion earthquake data to researchers and to the practicing

engineering community. The Task Committee focuses on instrumentation;

recording, processing, and analyzing strong-motion data; research to

ensure high-quality data dissemination; and analysis of ground motion

and structural motion in dynamic response to earthquakes.

2 . Scope of Work

The T/C coordinates strong-motion research, processing strong-motion

data, and disseminating information on the recorded behavior of

structures. In addition to regular exchange of data and publications,

the T/C's technical approach includes:

* Planning and conducting T/C workshops and meetings, held

generally in conjunction with UJNR Joint Meetings, often with

lASPEI/IUGG conferences and with World Conferences on

Earthquake Engineering.

* Conducting related technical sessions at professional society

meetings

.

* Creating procedures for the dissemination of significant

strong-motion digital data with regard for the rights and

expectations of (a) the owner, (b) the users of data, and (c)

the earthquake engineering community.

* Creating an up-to-date information exchange on computer

processing procedures and related research so all users of

processed data are aware of valid ranges and practical

limitations.

467
* Developing a research program for appropriate instrumentation

and dynamic analysis of structures with base isolation or

other response-reducing systems, under earthquake excitation.

* Developing a research program using temporary arrays and data

derived to study site response.

Accompl ishments

* Catalogs of strong-motion earthquake records observed both in

the United States and Japan are being exchanged. U.S. data

are published in the "Seismic Engineering Program Report", and

the Japanese data are published in "Strong-Motion Earthquake

Records in Japan". Many other publications were also

exchanged.

* Mr. Keiichi Tamura of PWRI is a Visiting Scholar at Stanford

University for one year from October 1989 working on strongmotion instrument arrays and data.

* Data tapes on the Loma Prieta earthquake of October 17, 1989,

processed by USGS and CDMG, were delivered to Japanese side.

* Disks containing data on the Chiba-Toho-Oki earthquake of

December 17, 1987, processed by the National Research Center

for Disaster Prevention, were delivered to U.S. side.

* The Japanese Technical Study Team for the Loma Prieta

Earthquake visited northern California, USA, between Nov. 16

& 29, 1989 to study the effects. The team and the U.S. hosts

included many panel members.

* Dr. Sasaki visited the U.S. during March 20-29, 1990, visiting

U.C Berkeley, USGS, Stanford, CALTRANS, CDMG; Loma Prieta

Earthquake effects in San Francisco, Oakland, and epicentral

area; and Washington (AID, NIST, NSF) . Panel members hosted

throughout; T/C A contacts included A. G. Brady and R. D.

Borcherdt, of USGS, and A. G. Franklin of USAEWES.

(Other Panel members included J. Gates, R. Hanson, H.S. Lew

and N. Raufaste)

.

Future Plans

* The T/C Chairmen will seek funding sources for a 3rd Workshop

on Processing Strong Motion Records, concentrating on array

data.

* Coordinate the analysis of records from arrays: downhole 3-D

arrays (geophysics and earthquake engineering) and structural

arrays (structural engineering)

.

468
* The T/C recognizes the efforts of the International Working

Group on the Effects of Surface Geology and encourages wider

participation by Panel members.

5. Information Exchange

* Will be performed within the domestic community through

technical meetings, conferences, seminars, and workshops.

* Will be performed through participation in post-earthquake

field investigations with concentration in areas where strongmotion stations have recorded data.

* Will be performed with the counterpart side through T/C

workshops; visitor exchanges, seminars, lectures; and

participation in annual Joint Panel meetings.

* The most crucial information exchange is rapid description of

recovered strong motion (and subsequently, digital strongmotion data on tape or disk) from significant earthquakes, for

example Loma Prieta Earthquake, CA of October 17, 1989.

6. Impact

* Most managers who process strong-motion recordings use

techniques based, or closely associated with, the work of this

T/C and its members. Foreign networks fitting this category

include:

AUSTRALIA - Australian Seismological Center, Bureau of Mineral

Resources, PO Box 378, Canberra ACT 2601.

CANADA - Pacific Geoscience Center, PO Box 6000, Sidney, BC.

V8L 4B2.

ITALY - The Italian National Commission for Research and

Development on Alternative Energy Sources (ENEA) , Rome.

MEXICO - Japan-Mexico Earthquake and Disaster Prevention

Center, Mexico City.

PERU - Japan-Peru Earthquake and Disaster Prevention Center,

Lima

.

TAIWAN - SMART Array

TURKEY - University of Bogazici, 80815 Bebek, Istanbul.

YUGOSLAVIA - Institute of Earthquake and Engineering

Seismology, Skopje.

469
* Use of seismological and strong-motion data, in convenient

form, is facilitated in countries other than where recorded.

* Dissemination of digital data from significant earthquakes

increases the research data base; application of this data in

design practice aids in reduction of earthquake hazards.

7 . Barriers

* Lack of adequate funding for T/C joint endeavors. '

* Establishing joint research efforts and maintaining close

interaction between participants are difficult unless visiting

research appointments are established. (Funding and staff

availability are often barriers)

.

470
Report of Task Committee on

(B) LARGE-SCALE TESTING PROGRAM

Date; May 16, 1990

Place National Institute of Standards and Technology

Gaithersburg, MD 20899

Attendees: Japan Side K.

S.

T.

S.

Ohtani (Chairman)

Okamoto

Kaminosono

Nakata

(NRCDP)

(BRI)

(BRI)

(BRI)

U.S. Side H.

T.

J.

S. Lew (Chairman)

Lew

Tissell (Observer)

(NIST)

(NCEL)

(American

Plywood Assoc.)

MISSION STATEMENT

1. Objective

The objective of the Task Committee is to develop performance data of

full-scale dynamic properties of structures to better resist wind and

seismic loads through both laboratory testing of prototype structures

and field testing of structures in situ. Improved full-scale test data

will validate the results of small-scale model tests and substantiate

the results of computer analyses of structural behavior.

2. Scope of Work

The Task Committee develops its research agenda in coordination with

other appropriate T/Cs, such as T/C's "C" and "D", for full-scale

evaluation of buildings, and other structures except bridges.

* Plans and conducts workshops and joint meetings to identify

research topics and develops joint research programs.

* Coordinates research projects carried out by various

laboratories in the U.S. and Japan. Facilitates publication

of research results and implementation of findings in codes

and standards.

* Facilitates exchange of research personnel, technical

information and available testing facilities.

471
3 . Accompl ishments

* As part of the second phase, analytical and experimental

research, including component and subassemblage tests, are

being carried out at several U.S. universities. Funding for

the third phase of the U.S. -Japan Coordinated Program for

Masonry Building Research has been postponed by the National

Science Foundation for U.S. researchers.

* The U. S. Technical Coordinating Committee on Masonry Research

was held in July 1989.

* The 5th Joint Technical Coordinating Committee on Masonry

Research (JTCCMAR) was held in October 1989 at Tsukuba, Japan.

Progress on individual research projects were presented by

principal investigators and future coordination of the Joint

Program was discussed.

* A preliminary Joint U.S. -Japan Precast Seismic Structural

Systems (PRESSS) Program Coordinating Meeting was held in

October 1989 in Japan.

* The first phase of the U. S. Precast Seismic Structural

Systems Program has been funded by NSF.

* The Government of Japan provided funding for the Japan PRESSS

Committee research program beginning April 1989. The Japan

Technical Coordinating Committee (TCC) Meeting was held in

April 1990, and it approved the scope of research of the

Japanese PRESSS project.

* Complement to the U.S. PRESSS program, NIST has carried out

research on precast concrete beam-column connections.

4 . Future Plans

* The 6th JTCCMAR will be held at Seattle, WA. in August 1990.

Immediately following this meeting the U.S. TCCMAR will be

held.

* The first Joint Technical Coordinating Committee (JTCC) on

PRESSS will be held in San Diego in November, 1990. The U.S.

TCC on PRESSS will meet prior to the first JTCC on PRESSS.

* A revised version of "Design Guidelines for Medium Rise

Reinforced Masonry Buildings" will be published in Japan in

August 1990.

* In support of IDNDR, findings of Joint research projects

carried out under the auspices of T/C "B" should be

disseminated widely to countries which has high seismic risks.

* Task Committee "B" recommends the establishment of a new Task

472
Committee on Passive and Active Energy Dissipation Systems.

* Continue exchange of information of the following topics

between appropriate U.S. and Japan organizations.

a. Seismic performance of composite and mixed construction,

b. Application of high strength construction materials to

structures in high seismic zones,

c. Testings of large-scale structures other than buildings

and bridges.

* Exchange of information on large-scale testing facilities and

large-scale testing programs should be encouraged.

Information Exchange

* Technical reports and research documentation were exchanged

with participating organizations in both countries.

Impact

* The results of the past joint research projects on reinforced

concrete and steel structures have been published in both U.S.

and Japanese technical journals. A number of key findings

have been incorporated into U.S. and Japanese building codes

and design guidelines.

* As a result of joint research projects, the state of practice

and understanding in earthquake engineering has been improved

in both countries. The results from the joint program have

further stimulated research in both countries.

* Facilitate joint research projects utilizing available testing

facilities in both countries.

Barriers

* There are no major technical barriers for the attainment of

the Task Committee mission. However, different building

practices and needs in the U.S. and Japan require careful

planning for effective joint research projects.

* Securing funding to support exchange of research personnel is

difficult in both countries. Funding agencies of both

countries should be informed about the technical benefits

resulting from the joint panel activities and encouraged to

participate in researcher exchanges and joint programs.

473
Report of Task Committee on

(C) REPAIR AND RETROFIT OF EXISTING STRUCTURES

Date

;

May 16, 1990

Place National Institute of Standards and Technology

Gaithersburg, MD 20899

Attendees: Japan Side S.

S.

T.

Nakata (Chairman)

Okamoto

Kaminosono

(BRI)

(BRI)

(BRI)

U. S. Side K.

G.

J.

G.

R.

E.

P. Chong (Chairman)

R. Fuller

R. Hill

E. Freeland

D. Marshall

Sabadell

(NSF)

(HUD)

(DOE)

(LLNL)

(NIST)

(NSF)

MISSION STATEMENT

1. Objective

The objective of the Task Committee is to encourage research and

provide technology transfer activities in materials, components, and

total buildings and structural systems related to evaluating the

performance of existing buildings and structures subjected to

environmental and earthquake forces. The research topics provide data

and information for the repair and retrofit of existing

buildings/structures

.

2. Scope-of-Work

The T/C work includes information exchanges and planning and hosting

workshops and seminars. Workshops are generally held in conjunction

with the annual Panel's meeting. They provide T/C members with an

opportunity to review related research and to direct plans for future

research, share and implement results. The T/C conducts related

technical sessions at professional society's meetings. The T/C studies

new materials and methods to accomplish repair and retrofit operations

on buildings e.g., fiberglass reinforced plastics, structural adhesives

and exploits use of automation and robotics for repair and retrofit.

Work focuses on coordinating research projects in Japan and the U.S. to

minimize duplication and to maximize benefits

474
3. Accompl ishments

* Held the 1-1/2 day NSF sponsored 5th Workshop on Evaluation,

Repair and Retrofit of Existing Structures prior to the 22nd

Joint Meeting, NIST, 1990 (24 delegates attended [12 from

Japan and 12 from U.S.] 19 manuscripts were presented [13 from

Japan and 6 from the US]).

* NSF issued a two-year initiative on Quantitative

Nondestructive Evaluation for Large Structural Systems in

April, 1990.

* NSF sponsored a workshop on non-destructive testing of civil

engineering structures at University of Southern California

(1988) .

* NSF issued a five-year initiative on Repair and Rehabilitation

Research for Seismic Resistance of Structures.

* A workshop was held at NIST in 1989 on International Workshop

on Sensors and Measurement Techniques for Assessing Structural

Performance

.

* FEMA published cost data for various methods of repairing and

retrofiting existing buildings.

* DOE has issued several design and evaluation guidelines for

mitigating the effects of earthquakes, winds and tornados.

* In the U.S., FEMA published ATC 21, "A Handbook for Seismic

Evaluation of Existing Buildings," and ATC 22, "Techniques for

Seismically Rehabilitating Existing Buildings" (drafts)

.

4. Future Plans

* A separate or joint (with T/C "D") workshop is being planned

on the Design Approaches and Construction Methods for Repair

and Retrofit of Buildings and Structures in May 1991 in Japan.

The workshop will focus on the use of newly developed methods

for repair, retrofit and evaluation of buildings/structures.

* In Japan, the Guidelines for Evaluating and Strengthening

Existing Buildings will be revised, and the Guidelines for

Evaluation of Earthquake Damage and Repair Methods of

Buildings will be published. Both guidelines will be

available after August, 1990.

* In support of IDNDR activities, T/C "C" would like to

participate in task committee activities related to repair and

retrofit of damages due to various disasters.

475
* A possible session on wind damages will be planned in 1992 in

U.S.

* NSF will sponsor a civil engineering non-destructive testing

workshop in Fall, 1990, in Boulder, Colorado.

Information Exchange

* Publications of the 5th Joint T/C "C" of "D" Workshops were

distributed, to all delegates and will be made available to

design professionals through technical presentations at

national meetings of professional societies.

* A U.S. Panel (Japanese Technology Evaluation Center Panel)

will be going to Japan in July, 90, to study and exchange

information on construction engineering.

* Published workshop reports and individual research papers;

they were exchanged with the U. S. and JAPAN-side members.

Impact

* Contributed to development of US and Japan design criteria for

repairing and strengthening reinforced concrete, steel, and

other buildings.

* T/C recommendations were incorporated into building codes and

professional practices, as illustrated by rehabilitation

programs used by the City of Los Angeles to upgrade existing

masonry buildings to resist earthquake forces.

* Influenced changes to U.S. and Japan design practices to

upgrade existing structures.

Barriers

* Greater progress needs be made to implement available research

results. Funding to sources need be identified to support

this endeavor. Funding agencies of both countries should be

encouraged to sponsor activities related to their mission, and

to support participants attendance at Joint Panel Meetings and

Workshops.

476
Report of Task Committee on

(D) EVALUATION OF STRUCTURAL PERFORMANCE

Date ; May 15, 1990

Place ; National Institute of Standards and Technology

Gaithersburg, MD 20899

Attendees ; Japan Side -

U.S. Side -

S. Okamoto (Chairman)

S. Nakata

T. Kaminosono

G. R. Fuller (Chairman)

K. P. Chong

E. Sabadell

J. R. Hill

R. D. Marshall

G. E. Freeland

J. Tissell (Observer)

(BRI)

(BRI)

(BRI)

(HUD)

(NSF)

(NSF)

(DOE)

(NIST)

(LLNL)

(Am. Plywood

Assoc .

)

MISSION STATEMENT

1 . Objective

The objectives of this Task Committee are to develop disaster

mitigation policies and programs to improve the capacity of existing

structures to resist wind and seismic forces. To provide adequate

performance evaluation, each country will coordinate development of

condition assessment, screening, and structural analysis methodologies.

Structures will then be analyzed and instrumented, and evaluated after

disasters

.

2. Scope of Work

* Develop a complete inventory of "benchmark" structures that

have been analyzed and instrumented for measurement of wind

and seismic response.

* Coordinate development of a uniform system of screening and

analyzing wind and seismic resistance and capacity of

structures in each country.

Develop sensor technology, instrumentation, and "expert"

systems to provide condition assessment of existing

structures

.

Evaluate masonry and precast concrete buildings, and seismicisolation systems, and include these findings in each

country's catalog of "benchmark" structures for post-disaster

performance evaluation.

477
* Plan and conduct workshops at future UJNR Joint Panel

Meetings, cooperatively with Task Committees "C" and "B".

3 . Accompl ishments

* A 1-1/2 day Joint Workshop with Task Committee "C" on "Repair,

Retrofit and Performance of Structures", was held at NIST on

May 12 and 14, 1990. Twenty-four delegates attended (12 from

Japan and 12 from the U.S.)* Nineteen manuscripts were

presented (13 from Japan-side and 6 from the U.S. side).

* Exchanged several reports on Post-earthquake and Posthurricane events: eg: Loma Prieta, and Hurricane Hugo, etc.

* Presented Applied Technology Council ATC-21 & ATC-22 draft

reports on "Evaluation Procedures for Existing Buildings." C.

Rojahn will supply copies to all workshop participants. (One

copy was given to Japan Side)

.

* Received Japanese MOC first and second year reports on stateof-art and catalog of buildings with seismic base isolation or

wind damping systems. The U.S. Side translated the documents

into English; they will be published by NIST and distributed

to Panel Members.

* A research workshop held in Washington, DC in August 1988

developed a U.S. research agenda for masonry. A working group

was related to structural performance of masonry buildings.

Evaluation techniques, nondestructive testing, and condition

assessment research was identified in the Final Report (copies

were provided to the Japanese Side)

.

* Copies of NIST SP 778 (ICSSC TRll) "Performance of Structures

During the Loma Prieta Earthquake of October 17, 1989" were

made available to all participants of the Workshop.

4 . Future Plans

* Develop a reporting system and data collection form to catalog

existing structures in the U.S. and Japan that have been

evaluated for wind and seismic resistance or have been

instrumented. Compile a computer data file for exchange

between countries and report its progress at each UJNR Joint

Panel Meeting.

* Encourage and enlist the participation of private industry,

consulting engineering firms, universities, and other

governmental agencies involved in the instrumentation,

evaluation, and condition assessment of existing structures.

* Encourage each country to develop sensor technology and

instrumentation of structures, for making condition

478
assessments and for determining the long term environmental

and aging effects on building components and materials.

* Participate in development of a new Panel effort on the

International Decade for Natural Disaster Reduction (IDNDR)

.

* Plan for Workshop prior to 23rd U.S. -Japan Panel Meeting in

Tsukuba, Japan. Japan Side will determine specific topics for

this Workshop. A Structural Performance Evaluation Workshop

on Wind Engineering is proposed to be held in the U.S. in

1992

.

Information Exchange

* Exchange methodologies for evaluation of structural

performance and reports of post disaster evaluations of

structures

.

* Provide most current data of "benchmark" structures that have

been evaluated and instrumented.

* Continue collection of related technical papers and other

literature and encourage development of an annotated

bibliography of information for periodic exchange.

* Prepare papers and reports for the 2 3rd UJNR Joint Panel

Meeting and associated "Workshop on Structural Performance

Evaluation"

.

* Japanese side will send to the U.S. T/C Chairman a copy of the

Final Revision of "Design Guidelines for Medium Rise

Reinforced Masonry Buildings" developed by TCCMAR-Japan (due

end of July 1990). U.S. Side will explore feasibility of

translating this manuscript into English. Summary report of

the design guidelines was printed in the Pre-Proceedings of

22nd Joint Meeting (3-3 Okamoto)

.

* Other papers of interest to Task Committee "D" in the PreProceeding of the 22nd Joint Panel Meeting are:

3-1 PRESSS Precast Concrete Joint Research Project (Okamoto) .

3- 2 U.S. Coordinated Program for Masonry Building Research

(Noland)

.

4- 2 Major Seismic Building Upgrades at Lawrence Livermore

National Laboratory (Freeland)

.

4- 9 Relation Between the Building Damage and Vibration

Properties of Ground (Ohtani)

.

5- 4 Performance of Wood-Framed Structures in Loma Prieta

479
Earthquake (Tissell)

.

* Copies of NISTIR 89-4153, "Sensors and Measurement Techniques

for Assessing Structural Performance," Proceedings of an

International Workshop held at NIST in Sept. 1988, were

distributed to all T/C members in Sept. of 1989.

6 . Impact

* Developed revised standards and building code provisions to

reduce destruction from high winds and earthquakes.

* Provided engineers with improved capability to conduct

condition assessment of existing structures for developing

cost effective repair, retrofit, strengthening or demolition

programs

.

* Increased the confidence level of the public, regulators,

policy makers and engineers, to provide assurance that major

structures will adequately resist high wind and earthquake

forces

.

7 . Barriers

* Difficulty in identifying priorities and inadequate resources

to document load carrying capacity of existing buildings

subjected to highly probable extreme winds and earthquakes.

* Limited resources for instrumenting buildings, for conducting

post-disaster evaluations, and for participating in foreign

travel for coordination of activities.

480
Report of Task Committee on

(E) NATURAL HAZARD ASSESSMENT AND MITIGATION THROUGH LAND USE

PROGRAMS

Date: May 15, 1990

Place ; National Institute of Standards and Technology

Gaithersburg, MD 20899

Japan Side - K. Kawashima (Chairman) (PWRI)

T. Iwasaki ( PWRI

)

K. Ohtani (NRCDP)

K. Tamura (PWRI)

T. Akagiri (GSI)

U.S. Side - W. Hays (Acting Chairman) (USGS)

H. Meyers (NOAA)

G. Brady (USGS)

MISSION STATEMENT

1 . Objective

The objectives of this Task Committee are to exchange information and

experiences on: 1) engineering characterization of ground motion 2)

mapping techniques for displaying ground motion data, 3) microzonation,

and 4) siting and design applications.

2. Scope of Work

New data acquired in Japan and the United States, and in other

countries (e.g., Chile, Mexico, and Armenia, USSR) will be analyzed for

future workshops in terms of the four above objectives.

3 . Accomplishments

Ideas for future activities were exchanged through correspondence from

T/C members of both sides.

4. Future Plan

* The Task Committee is exploring the feasibility of holding a

one or two day planning meeting in conjunction with the

anniversary meetings of the Loma Prieta earthquake which are

being planned for October and November 1990. The T/C meeting

could be held in California or in Tokyo in conjunction with

IDNDR meetings being planned in October.

* A conference on the first T/C objective, Engineering

Characterization of Ground Motion, planned for 1991, is

481
proposed to be held in conjunction with the 23rd Joint Panel

Meeting.

* In view of the increased ground motion data resulting from the

October 17, 1989, Loma Prieta earthquake and others throughout

the world, it was agreed that the Task Committee should focus

attention on increasing understanding of ground motion and

seismic design forces within the framework of its current

broad task. For such purpose, redefinition of the T/C name

as, "Ground Motion and Seismic Design Forces", is proposed.

5 . Information Exchange

Reprints of technical papers on ground motion, the Loma Prieta

Earthquake, and the IDNDR were exchanged at the 2 2nd Joint Panel

Meeting.

6 . Impact

The work of the T/C will be of great value to each country and

contribute to others throughout the world who have similar problems to

solve.

7 . Barriers

Lack of funding for travel to exchange information and to convene

workshops has been the principal barrier in the past.

482
Report of Task Committee on

(F) DISASTER PREVENTION METHODS FOR LIFELINES SYSTEMS

DateMay 15, 1990

Place National Institute of Standards and TechnologyAttendees: Japan Side K.

T.

Kawashima (Acting Chairman)

Iwasaki (PWRI)

( PWRI

)

U. S. Side S.

C.

R.

J.

T.

H.

K.

C. Liu (Chairman)

J. Astill

M. Chung (Observer)

Cooper

Lew

S . Lew

Sullivan

(NSF)

(NSF)

(NRC)

(FHWA)

(NCEL)

(NIST)

(FEMA)

MISSION STATEMENT

1 . Obi ective

The objectives of Task Committee "F" are to improve 1) the behavior of

lifeline systems during earthquakes and 2) engineering and other

seismic countermeasures such as damage estimation techniques and

inspection procedures. These technologies provide users with

improvements to existing standards of practice for safety and

seirviceability of lifeline systems. Lifeline systems such as gas, oil,

water, and sewage pipelines; and power, communication, and

transportation systems are crucial to the survival and health of a city

or community. Earthquakes affecting such systems cause severe social

and economic disruptions to communities and cause human suffering to

the residents.

2 . Scope of Work

* Plan and conduct T/C workshops and meetings; they generally

are held in conjunction with the annual UJNR Joint Meeting.

The purpose is to exchange the state-of-the-art knowledge and

practice and to identify cooperatively key opportunities for

exchange and studies which effectively implement research

results

.

* Conduct related technical sessions at professional society

meetings

.

* Create a research program on a selected lifeline system

(similar in concept to the Large-scale Testing Program) . For

example, through joint demonstration, produce criteria for

emergency design and loss evaluation and on-line monitoring

483
systems for lifeline operation.

* Conduct special activities such as translating, printing, and

distributing MOC PWRI ' s publication Manual for Repair Methods

for Civil Engineering Structures Damaged by Earthquake .

* Develop performance standards and a manual on repair,

restoration, and retrofit of lifelines.

Accomplishments

* A third UJNR Workshop on U.S. -Japan Lifeline Earthquake

Engineering was held in May 1989 in Tsukuba. The Workshop

proceedings were published and distributed to the research

communities in both countries.

* Researchers from both sides have contributed several technical

papers on lifeline systems at the 22nd joint panel meeting.

* Researchers from both sides have participated in the postevent field investigations on the collapse and damage of

lifeline structures in the San Francisco Bay area following

the Loma Prieta earthquake in October 1989. A number of indepth studies on earthquake behavior, safety, repair and

strengthening of the lifeline structures have been undertaken.

* Dr. Y. Sasaki, Japan Side T/C Chairman visited several

organizations including NSF and NIST in March 1990 to discus

active control research and to exchange information.

Future Plans

* Continue observation of instrumented buried pipes in

Parkfield, California, and Chiba, Japan and exchange data.

* Conduct the 4th Joint Workshop on Disaster Prevention for

Lifeline Systems tentatively scheduled in 1991 at an U.S.

location.

* During the 3rd Lifeline workshop in May 1989, the importance

of active control research was identified for mitigation of

damage to lifeline structures during earthquakes. T/C "F"

recommends establishment of a new task committee on

Active/Passive control systems for exchanging technical

information, undertaking joint research, and coordinating

research programs carried out by panel member organizations.

* Initiate a new cooperative research project on seismic control

systems for lifelines to mitigate damage. This work would

complement the above activity.

484
5 . Information Exchange

* Domestic community: through technical meetings, conference

sessions, seminars, and workshops.

* Between U.S. and Japan: through T/C workshops and annual

Joint Panel Meetings.

* Visits by researchers and participation in technical meetings

sponsored by organizations such as NCEER and EERI.

* Short-term (3 to 12 months) researcher exchanges to

participate in on-going lifeline research such as performed at

the University of Michigan and PCA in the U.S. and at BRI in

Japan.

6. Impact

* Contributed to increased field engineers' technical knowledge

on repair methods, through seminars with lectures by T/C

members using the MOC Manual in nine major cities in Japan,

including Tokyo and Osaka; more than two thousand participants

attended seminars.

* Contributed to the assessment of design, construction, and

operating standards for lifeline systems through T/C member

participation in the committee work e.g., the Technical

Council on Lifeline Earthquake Engineering (TCLEE/ASCE) and

the U. S. Building Seismic Safety Council.

7 . Barriers

* Insufficient funds to conduct annual Task Committee Workshops

and to publish proceedings and reports.

* Lack of participation by U.S. industrial organizations and

utilities in research or in implementation of findings due to

a reluctance to adopt new technologies and practices.

485

I

I
Report of Task Committee on

(H) SOIL BEHAVIOR AND STABILITY DURING EARTHQUAKES

Date : May 16, 1990

Place ; National Institute of Standards and Technology

Gaithersburg, MD 20899

Attendees: Japan Side T. Iwasaki (Acting Chairman)

T. Uwabe

(PWRI)

( PHRI

)

U.S. Side - A. G. Franklin (Chairman)

R. L. Ledbetter

W. E. Roper

C. J. Astill

(WES)

(WES)

(COE)

(NSF)

MISSION STATEMENT

1 . Objective

The objectives of this Task Committee are to enhance technology for:

predicting the dynamic behavior of soils, establishing procedures to

analyze dynamic soil-structure interaction, and modifying the dynamic

behavior of foundations and earth structures in order to assure their

safe performance during earthquakes. Government agencies with

responsibility for public works have a duty to provide for public

safety and to make economical use of public funds. Consequently,

government agencies have a need for more accurate methods of predicting

soil behavior, and for improved, economical methods of controlling soil

behavior, to assure the seismic safety of public works and to minimize

the costs of construction by avoiding unnecessary over-design.

2 . Scope of Work

* Present technical papers at annual joint panel meetings on

technological developments and the state-of-the-art and

practice related to soil behavior and stability during

earthquakes.

* Exchange information and technical data relating to field

performance, research and methods of practice.

* Plan and conduct T/C workshops in coordination with proposed

or ongoing cooperative research programs.

* The Task Committee plans and conducts programs of cooperative

research, exchange of researchers between U.S. and Japanese

government laboratories, publication of research results and

recommended practice. The benefits realized from available

research funds can be amplified by bringing greater resources

to bear on problems of mutual interest, and by coordinating

486
research tasks to make the best use of the respective

strengths of the two sides' research capabilities.

* Exchange visiting researchers using mechanisms provided by

formal and informal cooperative research programs. Such

exchanges promote technical information exchanges,

familiarization with methods of practice on the respective

sides, and mutual understanding and cooperative relationships

between researchers on the two sides.

Accomplishments

* U.S. researchers presented 5 papers, and Japanese researchers

presented 11 papers, on soil behavior and stability during

earthquakes, at the 21st and 2 2nd joint panel meetings

5/89, 5/90

* Proceedings of the Workshop on the In-Situ Testing, held at

San Francisco, California in 1985 was published and

disseminated to the Panel members 4/89

* A planning meeting for cooperative research on preventive

measures against soil liquefaction was held at Tsukuba prior

to the 21st joint meeting .....5/89

* Mr. Keiichi Tamura of the PWRI is a one-year Guest Researcher

at Stanford University, working on Seismic Response

Characteristics of Ground 10/89

* Mr. Motoki Kazama of the PHRI is at the Scripps Institute of

Oceanography as a one-year Guest Researcher, working on

Effects of Offshore Environments on Seismic Response of Soil

Deposits and Pile Foundations 10/89

* Mr. Nario Yasuda of the PWRI completed a one-year Guest

Researcher assignment at the U.S. Bureau of Reclamation, doing

research in Seismic Behavior of Soils ............ 1/90

* Dr. Yasushi Sasaki of the PWRI visited sites of damage around

the San Francisco Bay Area and the epicentral area of the Loma

Prieta Earthquake. He was assisted by several U.S. Panel

Members, including the T/C (H) U.S. Chairman 3/90

* A set of 3 5mm slides showing the effects of the Loma Prieta

Earthquake was provided to the Japanese side by the U.S. Side

3/90

* Dr. Ashok K. Chugh, of the U.S. Bureau of Reclamation, visited

the PWRI in March and April 1990, to study the Dynamic

Behavior of Rock-Fill Dams 4/90

487
Future Plans;

* Identification of research tasks on preventive measures

against soil liquefaction that are of common interest, in

accordance with the results of the planning meeting held at

Tsukuba in May 1989 9/90

* Hold the Second U.S. -Japan Workshop on Remedial Treatment of

Potentially Liquefiable Soils, in Japan at a time to be

determined between October 1990 and March 1991 ...2/91

* Develop plans for a cooperative research program to verify

soil-pile interaction models for marine structures subject to

earthquakes, measurement of sea floor earthquake motions, and

the stability of gravity-based structures during seismic

events c 12/89

* T/C (H) , at the 22nd Joint Panel Meeting, discussed the need

for technological innovation and research to assist in saving

lives following natural disasters caused by earthquakes. The

principal areas of need were identified as improved technology

for locating survivors, and the development of new equipment

and systems for safe excavation of debris to remove survivors

from collapsed structures. T/C (H) recognizes this subject is

outside their area of responsibility but feels it is important

and should be discussed by the full Joint Panel. The Task

Committee suggests the following options for implementation:

(a) Establish a new Task Committee which will address this

topic area and provide the focal point to share experience and

research in this area and (b) Incorporate this topic into

another T/C e.g., T/C (D) : on Evaluation of Structural

Performance. T/C (H) will continue to explore methods whereby

this topic will receive appropriate visibility.

Information Exchange :

* T/C workshops, exchange of visitors, and participation in

annual joint panel meetings, and presenting technical papers.

* Exchange of researchers between U.S. and Japanese government

laboratories

.

* Visits by researchers and participation in technical

activities such as post-earthquake field investigations.

* Professional community in the U.S. and Japan; publication of

papers in journals; publication of research reports by the

respective agencies; participation in professional society

meetings and conferences; seminars; and workshops.

488
6 . Impact

Through the exchange of researchers and the diffusion of earthquakerelated technical data, experience, and information on methods of

practice, the state of technology in geotechnical earthquake

engineering has been raised on both sides. Through the cooperative

research program on in-situ testing of soils which was carried out

after the 1983 Nihonkai-Chubu Earthquake, improved and more accurate

methods of using in-situ tests for evaluating the liquefaction

potential of soils were achieved; these have been in general use in the

United States since 1985. The Loma Prieta Earthquake stimulated

renewed awareness of the need for further research in soil liquefaction

and amplification effects of soft soils.

7 . Barriers

* Insufficient U. S, and Japan Government funds allocated to

adequately address earthquake engineering research.

* Administrative limitations on foreign travel.

489
Report of Task Committee on

(I) STORM SURGE AND TSUNAMI

Date : May 15, 1990

Place ; National Institute of Standards and Technology

Gaithersburg, MD 20899

Attendees : Japan Side - K. Yokoyama (Temp. Chairman) (PWRI)

H. Ito (MRI)

U.S. Side - H. Meyers (Chairman) (NOAA)

C. Barrientos (NOAA)

W. E. Roper (COE)

MISSION STATEMENT

1 . Objective

The objective of this T/C is to mitigate damages from storm surges and

tsunamis through shared technologies, research, information sharing,

and cooperative work. Storm surge and tsunami are hazards capable of

inflecting damage of disastrous proportions. Storm surges are

associated with hurricanes and typhoons where high winds and the

mounding of water under the low barometric pressure of the storm's eye

can cause extensive flooding and severe wave action. Tsunamis are

predominately caused by underwater earthquakes and, to a lesser extent

volcanic activity. Depending on the distance from the source tsunamis

arrive within minutes to up to a day after generation.

2 . Scope of Work

* Exchange results of research on storm surge and tsunami

occurrence, generation, propagation, and costal effects. This

includes observations on historical, current, and theoretical

tsunamis. Of particular interest is the effort by US and

JAPAN to acquire deep ocean tsunami measurements.

* Exchange results and status of anti-storm surge and tsunami;

activities including analysis of the problem, planning,

warning, and engineering approaches.

* Exchange information on planned and ongoing projects relating

to storm surge and tsunamis.

* Exchange information on development of technologies such as

computer programs to predict travel times, land-fall

locations, run-up heights, and wave characteristics and

analysis; improved instrumentation, and use of satellite

communications for detection and warning.

490
* Facilitate dissemination through exchange of literature,

technical reports at joint meetings, special workshops, joint

projects, and direct interaction among participants.

Accomplishments

* Exchanged digital bathymetric data with the Japanese side; US

side received a computer program to calculate tsunami travel

time charts.

* Exchanged computer program forecasting storm surge run-up.

* Exchanged technical reports on related subjects.

* Expanded the number of active U.S. participants with

representation from COE, University of Colorado, and NOAA, and

NSF.

* The U.S. side has compiled detailed information on all

tsunamis affecting the U.S. and its possessions. The

publication, "United States Tsunamis 1690-1989" has been

printed and given to the Japanese side.

Future Plans

* Develop exchanges of data and information on activities

related to deep ocean tsunami measurements.

* A Second Tsunami Workshop is being arranged in Hawaii,

November 5-7, 1990 to address advances since the first

Workshop held in 1983 and to bring together computer modelers

and potential users of modeling results. There will be

approximately 8 participants from the Japanese side and 10

from the U.S. side. Site visits are also being planned.

* Explore and promote joint undertakings such as instrument

testing, joint publications, scientist exchange, satellite

communications, and assistance in contacts within the US and

JAPAN

.

* NOAA has adopted the travel time program developed by Mr.

Okada (MRI/JMA) using NOAA gridded bathymetric data to run on

NOAA computers. The first ever computed travel time charts

have been prepared for the Caribbean Sea using this program

developed by Mr. Okada. A copy of the published paper will be

sent to the Japanese side.

* The U.S. and Japanese sides agree that there is a need to hold

workshops at 2 or 3 year intervals to exchange the latest

research and technology relating to various aspects of storm

surge and tsunami hazards.

491
* Continue exchange of publications, bibliographic data,

technical reports, and personal contacts.

6 . Impact

* Increased cooperation among scientists and engineers working

in these fields.

* Hasten implementation of new modeling and design techniques

developed by the US and Japan.

* Facilitated wider dissemination of information and technology

to scientists and engineers in both countries.

7 . Barriers

* Insufficient funding from government and private sector

organizations to participate in T/C workshops and the Panel's

annual joint meetings.

* The tsunami problem is less severe in the US than in JAPAN

which leads to a smaller community of researchers. There are

few US researchers focusing on tsunamis at any specific

location; thus highlighting the importance of T/C periodic

meetings and workshops.

* There are different approaches and degree of significance in

dealing with problems in the U.S. and JAPAN e.g., in U.S. much

of the hazard is from remote sources which leads the

government to emphasizing developments of warning and

evacuation schemes while in JAPAN the engineering approach is

needed due to more locally generated tsunamis which often do

not provide sufficient times for adequate warning and

evacuation. This barrier is partially being addressed by

inviting engineers and non-government specialists to

participate as temporary or permanent members for the U.S.

side at the annual meetings.

492
Report of Task Committee on

(J) WIND AND EARTHQUAKE ENGINEERING FOR TRANSPORTATION SYSTEMS

Date ; May 15, 1990

Place ; National Institute of Standards and Technology

Gaithersburg, MD 20899

Attendees; Japan Side - K. Yokoyama (Acting Chairman) (PWRI)

K. Hasegawa (PWRI)

U.S. Side - J. D. Cooper (Chairman) (FHWA)

MISSION STATEMENT

1 . Objectives

The objectives of this Task Committee are to plan, promote, and foster

research on the behavior of highway bridges when subjected to wind and

seismic forces and to disseminate research results and provide

specifications and guidelines based on the task committee's findings.

Surface transportation systems for movement of goods and people play a

vital part of commerce and intercourse between people. Highway bridges

are especially influenced by the forces of wind and earthquakes because

of their open exposure to those forces.

2. Scope-of-Work

* The scope of the work is applicable mostly to highway bridges

without any limitation on their size and function: such as

existing bridges and new bridge designs; whole system of

bridges; and/or to single components of a bridge.

* The mission is performed through; conducting workshops,

exchanging researchers, developing methods for design

evaluations and test procedures, inspection techniques,

rehabilitation and maintenance specifications and policies,

and performing cooperative research programs and other

relevant cooperative administrative activities.

3. Accompl ishments

* United States and Japanese post earthquake investigative teams

collaborated to investigate the damage to transportation

systems with emphasis on highway bridges caused by the Loma

Prieta Earthquake of October 17, 1989. Technical information

was exchanged at the 2 2nd Joint Panel Meeting.

* Conducted coordinated research studies on the seismic

performance of bridge piers and columns. Details of this work

493
involved determining the performance of reinforced concrete

piers and columns subjected to dynamic cyclic loading,

performing model tests on the failure of reinforced concrete

piers, testing full-scale concrete columns and for the

behavior of concrete filled steel tubes.

* Held the sixth bridge workshop in Lake Tahoe, Nevada U.S. in

1990 and the fifth bridge workshop in Tsukuba, Japan in 1989,

dealing with the determination and evaluation of the

performance and strengthening of bridge structures, structural

monitoring, repairs, safety and non-destructive evaluation.

Special attention was given to the seismic and wind loadings

of modern cable-stayed bridges, vibration suppression, base

isolation and dynamic control techniques.

* Both sides presented technical papers at the 21st and 22nd

UJNR Joint Meeting and participated in technical study tours

to buildings, bridges, factories, construction sites, and

instrumented structures.

Future Plans

* Conduct the 7th bridge workshop just prior to the 23rd Joint

Panel meeting in Japan.

* T/C (J) recognizes the importance of developing active and

passive control systems to mitigate earthquake and wind

induced damage to bridges. Because of the potential

application and impact of these control systems to mitigate

damage on other types of structures, T/C (J) recommends the

establishment of a new Panel T/C to review the feasibility for

and encourage and support the development of active and

passive control systems to reduce earthquake and wind induced

structural damage.

* Investigate and exchange technical information on improved

seismic retrofit and strengthening procedures for highway

bridges based on experimental, analytical, and field studies.

* Continue experimental research study on the seismic

performance of bridge piers and columns, and continue emphasis

on base isolation.

* Develop a coordinated research study on seismic, aeroelastic,

and aerodynamic response of cable-supported bridges with

emphasis on cable inspection, vibration and corrosion

protection.

* Develop a coordinated research study to compare the seismic

design criteria for bridges in Japan and the U.S., including

application of limit state design.

* Encourage a coordinated research study on seismic response

494
control, system identification techniques, and non-destructive

evaluation of bridge structures.

5 . Information exchange

* Exchanged technical reports, research program documentation,

construction logs, design plans, and assorted photographs and

TV videos.

* Mr. Keiichi Tamura from PWRI studied at Stanford University

and Mr. Kinj i Hasegawa from PWRI studied at the University of

California,

Berkeley. Mr. S. Unjoh returned to PWRI in March 1990 after

working for one year at the University of Southern California.

6 . Impact

* Greater uniformity was achieved in wind engineering test

procedures and modeling resulting in more efficient solutions

to the aerodynamic bridge problem.

* The workshops and study tours in the two countries have

facilitated technology transfer, as an example, in the area of

cable protection and vibration suppression of cable-stayed

bridges.

* With respect to the earthquake protection of bridges, the free

exchange of literature, instrumentation technology, and

earthquake response data has led to the advancement in design

technology, and development of preliminary retrofit

guidelines, and to the calibration and verification of

specifications.

* A set of priorities for needed research from a more

international viewpoint has reduced the conduct of duplicative

research efforts.

7 . Barriers

There are no major technical barriers to achieve the Task Committee

mission. However, financial barriers, especially funding needed to

participate in workshops and panel meetings and for conducting actual

bridge tests and large scale model tests, limit the effectiveness of

the work in this Task Committee's work.

495

NIST-114A U.S. DEPARTMENT OF COMMERCE

(REV. 3-89) NATIONAL INSTITUTE OF STANDARDS AND TECHNOLOGY

BIBLIOGRAPHIC DATA SHEET

1. PUBLICATION OR REPORT NUMBER

NIST/SP-796

2. PERFORMING ORGANi2^TION REPORT NUMBER

3. PUBUCATION DATE September 1990

4. TITLE AND SUBTITLE Wind and Seismic Effects —Proceedings of the 22nd Joint Meeting of the U.S. -Japan

Cooperative Program in Natural Resources Panel on Wind and Seismic Effects

5. AUTHOR (S) Noel J. Raufaste, Editor

6. PERFORMING ORGANIZATION (IF JOINT OR OTHER THAN NIST, SEE INSTRUCTIONS)

U.S. DEPARTMENT OF COMMERCE

NATIONAL INSTITUTE OF STANDARDS AND TECHNOLOGY

GAITHERSBURQ, MD 20899

7. CONTRACT/GRANT NUMBER

8. TYPE OF REPORT AND PERIOD COVERED

Final

I 9. SPONSORING ORGANIZATION NAME AND COMPLETE ADDRESS (STREET, CITY, STATE, ZIP)

SAME AS ITEM // 6 ABOVE.

10. SUPPLEMENTARY NOTES

DOCUMENT DESCRIBES A COMPUTER PROGRAM; SF-185, FIPS SOFTWARE SUMMARY, IS ATTACHED.

11. ABSTRACT (A 200-WORD OR LESS FACTUAL SUMMARY OF MOST SIGNIFICANT INFORMATION. IF DOCUMENT INCLUDES A SIGNIFICANT BIBLIOGRAPHY OR UTERATURE SURVEY, MENTION IT HERE.) The 22nd Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects was held at the

National Institute of Standards and Technology from May 15-18, 1990. This publication, the

proceedings of the Joint Meeting, includes the program, list of members, panel resolutions,

task committee reports, and 39 technical papers.

The papers were presented under six themes: (I) - Wind Engineering, (II) - Storm Surge and

Tsunamis, (III) - Joint Cooperative Research Program, (IV) - Earthquake Engineering, (V) - Loma Prieta Earthquake, and (VI) - Summaries of Task Committee Workshop Reports.

12. KEY WORDS (6 TO 12 ENTRIES; ALPHABETICAL ORDER; CAPITALIZE ONLY PROPER NAMES; AND SEPARATE KEYWORDS BY SEMICOLONS)

accelerograph; bridges; codes; concrete; design criteria; disaster; earthquakes; geotechnical

engineering; ground failures; inelastic ; lifelines; liquefaction; Loma Prieta; masonry;

repair and retrofit; risk assessment; seismicity; soils; standards; storm surge; structural

engineering; tsunami; and wind loads.

13. AVAILABIUTY UNUMITED

FOR OFFICIAL DISTRIBUTION. DO NOT RELEASE TO NATIONAL TECHNICAL INFORMATION SERVICE (NTIS).

14. NUMBER OF PRINTED PAGES

ORDER FROM SUPERINTENDENT OF DOCUMENTS, U.S. GOVERNMENT PRINTING OFFICE, WASHINGTON, DC 20402. ORDER FROM NATIONAL TECHNICAL INFORMATION SERVICE (NTIS), SPRINGFIELD. VA 22161.

524

IS. PRICE

ELECTRONIC FORM U.S. GOVERNMENT PRINTING Of FICE: 1990— 261-913 /20707
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