If"

Wind and Seismic Effects

Proceedings of the 32nd Joint Meeting

NIST Special Publication 963

Stephen A. Cauffman, Editor

QC

100

.U57

No. 963

2001

C.2

Monal Institute of Standards and Technology

inology Administration, U.S. Department of Commerce N T E N N I
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'At Boulder, CO 80303. ^Some elements at Boulder, CO.
Wind and

Seismic

Effects

NIST SP 963

PROCEEDINGS OF

THE 32ND JOINT

MEETING OF

THE U.S.-JAPAN

COOPERATIVE PROGRAM

IN NATURAL RESOURCES

PANEL ON WIND AND

SEISMIC EFFECTS

Issued April 2001

Stephen A. Cauffman,

Structures Division

Building and Fire Researcii Laboratory

Gaithersburg, MD 20899-8610

U.S. Department of Commerce

Donald L. Evans, Secretary

EDITOR

National Institute of Standards and Technology

Karen H. Brown, Acting Director
Certain commercial entities, equipment, or materials may be identified in this document in order to describe an experimental procedure or concept adequately. Such

identification is not intended to imply recommendation or endorsement by the National Institute of Standards and Technology, nor is it intended to imply that the

entities, materials, or equipment are necessarily the best available for the purpose.

National Institute of Standards and Technology Special Publication 963

Natl. Inst. Stand. Technol. Spec. Publ. 963, 549 pages (April 2001)

CODEN: NSPUE2

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PREFACE

This publication contains the proceedings of the 32" Joint Meeting of the U.S. -Japan

Panel on Wind and Seismic Effects. The meeting was held at the National Institute of

Standards and Technology (NIST), Gaithersburg, Maryland (USA), during May 16-19,

2000. The proceedings include the program, list of members, task committee reports, and

technical papers submitted to the Joint Panel Meeting, and a summary of the technical

site visits. BACKGROUND

Responding to the need for improving engineering and scientific practices through

exchange of technical data and information, research personnel, and research equipment,

the United States and Japan in 1961 created the U.S. -Japan Cooperative Science

Program. Three collateral programs comprise the Cooperative Science Program. One of

the three, the U.S.-Japan Cooperative Program in Natural Resources (UJNR), was created

in January 1964. The objective of UJNR is to exchange information on research results

and exchange scientists and engineers in the area of natural resources for the benefit of

both countries. UJNR is composed of 18 Panels, each of which is responsible for specific

technical subjects.

The Panel on Wind and Seismic Effects was established in 1969. Seventeen U.S. and

seven Japanese agencies participate with representatives of private sector organizations to develop and exchange technologies aimed at reducing damage from high winds,

earthquakes, storm surge, and tsunamis. This work is produced through collaboration

between U.S. and Japanese member researchers working in eleven task committees.

Each task committee focuses on specific technical issues, e.g., earthquake strong motion

data. The Panel provides the vehicle to exchange technical data and information on

design and construction of civil engineering lifelines, buildings, and waterfront

structures, and to exchange high wind and seismic measurement records. Annual

meetings alternate between Japan and the U.S. (odd numbered years in Japan; even numbered years in the U.S.). These one-week technical meetings provide a forum to discuss ongoing research and research results; they are followed by a series of technical

site visits.

The National Institute of Standards and Technology (NIST) provides the U.S. -side

chairman and secretary-general. The Public Works Research Institute (PWRI) provides

the Japan-side chairman and secretary-general.

Annual meetings provide the mechanism for interaction between U.S. and Japanese

researchers in wind and seismic engineering, creating opportunities to gain valuable

information and to engage in cooperative research. Through these opportunities, the

Panel member organizations have realized important advances in building and structures technology. The Panel's activities have resulted in improved building and structures codes and standards in both countries; for example, the Panel:

iii
• created and exchanged digitized earthquake records used as the basis of design and

research for Japan and the United States; • transferred earthquake engineering information and strong-motion measurement

techniques for use by seismically active countries, e.g., Australia, Canada, Italy,

Mexico, Peru, Taiwan, Turkey, and North Africa;

• produced data that advanced the design and construction of bridge structures;

• produced large-scale testing data that advanced the seismic design standards for

buildings;

• improved in-situ measurement methods for soil liquefaction and stability under

seismic loads;

• created a database on storm surge and tsunamis and verified mathematical models of

tsunami and storm surge warning systems;

• established a library resource of current research on wind and earthquake engineering

and on storm surge and tsunamis;

• published proceedings of Panel Meeting and Task Committee Workshops;

• provided earthquake science and technology expertise to the U.S.-Japan Natural

Disaster Reduction Initiative of the U.S.-Japan Framework for New Economic

Partnership (Common Agenda);

• created a web site for promoting the Panel's activities widely.

The Panel's efforts are exemplary of effective joint research and technology delivery

between researchers from the U.S. and Japan. Since its creation, more than 1900 papers

have been presented in 32 Joint Meetings, more than 60 Task Committee Workshops

were held, and over 200 guest researchers have participated in exchange programs. The

Panel provides important information about the U.S. and Japan's civil engineering

thrusts, which influence both countries' research and provide the basis for improvements

in building and structures codes and standards.

HIGHLIGHTS OF THE 32"'' JOINT PANEL MEETING

• Technical presentations:

Forty-four papers were presented during ten sessions.

Important knowledge was exchanged for research and practice, e.g., information

systems for risk assessment and disaster response using Geographical Information

Systems; performance-based design of buildings, bridges, and port facilities; wind

and earthquake engineering for buildings, bridges and other structures; research

findings and lessons learned from recent natural disasters; and research findings on public health issues following natural disasters.

• Task Committee activities:

The T/C structure is an extremely effective vehicle to explore in depth the

technologies and methods being used by both countries.

Five Task Committee Workshops and Meetings were held in the past year.

Five Task Committee Workshops/Committee Meetings are planned for the

coming year.

iv
• Strategic Plan Development:

Recognizing its strengths and accomplishments and a desire to position itself functionally

and organizationally to address future challenges, the Panel agreed to appoint a Joint AdHoc Planning Committee to develop a strategic plan for the Panel. The Joint Ad-Hoc

Planning Committee will:

Evaluate the current organization, task committee structure, annual technical

meeting and site visit program, cooperative research projects, and related

operating procedures.

Assess current and emerging technical needs and interests of government agencies

in the two countries.

Review options for streamlining and increasing the efficiency of Panel operations

and for moving toward more effective operating procedures.

Develop a joint strategic plan for the Panel and present it to the Panel for adoption

in early 2001.

John L. Gross, Secretary-General

U.S.-side Panel on Wind and Seismic Effects

V

CONTENTS

PREFACE iii AGENDA OF JOINT MEETING AND TECHNICAL SITE VISITS 7

LIST OF PANEL MEMBERS 17

RESOLUTIONS 35

PAPERS 41

THEME 1. Information Technology for Land Management and Natural Disaster

Prevention Countermeasures

Web-Based Geographic Information System Support to Disaster Response

William E. Roper (NIST) 45

Draft Manual For Seismic Information System Development

Hideki SUGITA and Tomofumi NOZAKI (PWRI) 63

GIS and Pre- and Post-Disaster Risk Assessment

Stuart Nishenko and Claire Drury (FEMA)

Scott Lawson, Thomas Durham, and Jawhar Bouabid (Durham Technologies) 73

Early Estimation System for Tsunami Hazard

KIRIYAMA Takaharu and OKAYAMA Kazuo (National Land Agency) 83

THEME 2. Wind Engineering

Rain/Wind Induced Vibration of Bridge Cables in the United States

Harold R. Bosch (FHWA) 97

Study on Performance-Based Design Methods for Wind Effects on Highway Bridges

Hiroshi SATO and Ken-ichi OGI (PWRI) 107

Weather Situation on Ground Observed at Fire Stations - Case of Typhoon Vicki

Yasuo OKUDA and Hisashi OKADA (BRI)

Hirohiko ISHIKAWA (Kyoto University)

Takeshi FUJII (Kyoto Sangyo University) 115

THEME 3. Earthquake Engineering

Seismic Testing of a 1/20-Scale 2D Model of Konya Dam

Robert L. Hall and Mostafiz R. Chowdhury (US Army Eng. R&D Center)

Enrique E. Matheu (Louisiana State University) 131

1
Development of Smart Systems for Building Structures

Shunsuke OTANI (University of Tokyo)

Hisahiro HIRAISHI, Mitumasa MIDORIKAWA, Masaomi TESHIGAWARA,

Hideo FUJITANI, and Taiki SAITO (BRI) 141

Nonlinear Rock Mechanics Analysis of Gravity Dams Under Combined Flood and

Earthquake Loading

Luis A. de Bejar and Robert L. Hall (US Army Eng. R&D Center) 155

Outline and Pre-Analytical Study on Damage Detection Tests of Five-Story Steel Frame

With Simulated Damages

Masaomi TESHIGAWARA, Takashi KAMINOSONO, Mizuo INUKAI, Hiroshi,

ISODA, Klichi MORITA, and Kazuya NOGUCHI (BRI) 167

The Influence of High Confining Stress on the Cyclic Behavior of Saturated Sand

R. Scott Steedman, Richard L. Ledbetter, and Mary E. Hynes (Waterways Experiment

Station) 173

Soil Amplification Factor for Seismic Design of Buildings

Izuru OKAWA, Masanori IIBA, Mitsumasa MIDORIKAWA, and

Shin KOYAMA (BRI)

Kenji MIURA (Kajima Corporation) 195

Construction of A Base-Isolated House for Observation of Isolation Effects During

Earthquake and Wind

Masanori IIBA, Mitsumasa MIDORIKAWA, and Yasuyuki YAMANOUCHI (BRI)

Massayoshi IKENAGA (Oiles Corporation)

Kenichi MACHIDA (Sumitomo Forestry Co., Ltd.) 203

Design Procedures for Structures with Supplemental Dampers

Fahim Sadek and Michael A. Riley (NIST) 213

Seismic Design Ground Motions for Strait-Crossing Projects in Japan

Masahiko YASUDA, Keiichi TAMURA, Shojiro KATAOKA

and Yoshihiro NAKAO (PWRI) 223

Development of Seismic Isolation Technology For Underground Structures and the

Application

Shigeki UNJOH, Junichi HOSHIKUMA, and Kazuhiro NAGAYA (PWRI)

Takeyasu SUZUKI (Technical Institute of Kumagai-gumi Corp.) 239

Earthquake Motions at an Embankment Dam Base and an Estimadon Method of Incident

Seismic Waves Using the Observations

Yoshikazu YAMAGUCHI and Tomoya IWASHITA (PWRI) 253

Reproduction of Large-Scale IG Test on Dry Sand Deposits and Pile Foundations Using

Centrifuge Modeling (Step-2)

Chikahiro MINOWA (National Research Center for Earth Science & Disaster

Prevention)

Masayoshi SATO (PWRI)

Takaaki KAGAWA (Wayne State University)

Akio ABE (Tokyo Soil Research Co.) 263

2
An Experimental Study for the Critical State of Footing Loaded Moment and Shear

By Tension

Shoichi SAEKI (PWRC)

Jiro FUKUI, Masahiro SHIRATO, and Shin-ichiro FURUSHO (PWRI) 271

Dynamic Characteristics of Magneto-Rheological Damper

Hideo FUJITANI (BRI)

Katsuaki SUNAKODA, Hiroshi SODEYAMA (Sanwa Tekki Corp.)

Norio IWATA, Satsuya SODA (Waseda University) 283

THEME 4. Performance Based Design Methods

Comparison of Performance Required by Bridge Design Codes in Various Countries

Michio OKAHARA, Jiro FUKUI, and Masahiro NISHITANI (PWRI)

Kenji MATSUI (CTI Engineering Co., Ltd.) 297

Recommended Changes to the AASHTO Specifications for the Seismic Design of

Highway Bridges (NCHRP Project 12-49)

Ian M. Friedland (ATC)

Ronald L. Mayes (Bridge Eng. Consultant) 313

Earthquake Resistant Design of Port Facilities

Takashi NAGAO and Atsushi NOZU (Port and Harbour Research Institute) 325

NIST Research on Structural Performance of Housing Systems

Fahim Sadek and Michael A. Riley (NIST) 333

Performance-Based Damage Assessment and Design Verification of U.S. Housing

in Extreme Earthquake and Hurricane Events

Jay H. Crandall (NAHB)

William Freebome (HUD) 343

American Lifelines Alliance - Reducing Risk to Utility and Transportation Systems

from Natural Hazards

Stuart Nishenko and Tim Sheckler (FEMA)

William U. Savage (FEMA and Pacific Gas and Electric Co.)

Douglas Honegger (Arroyo Grande, CA) 363

THEMES. Public Health

Public Health Assessments in Post-Disaster Settings: Recommendations for a

Multidiscipiinary Approach

Josephine Malilay (CDC) 369

Assessment Models For Earthquake Fatalities and an Interpretation of The Deaths in The

1999 Kocaeli, Turkey Earthquake

Keishi SHIONO (Nagaoka College of Technology) 377

Predictors For People's Response to a Tornado Warning, Arkansas, March 1, 1977

Lina Balluz, Laura Schieve, Talmage Holmes, Stephanie Kiezak, and

Josephine Malilay (CDC) 391

3
Post-Disaster Carbon Monoxide Poisoning: An Emergency Health Problem

W. Randolph Daley (CDC) 395

Assessment of Public Health Needs Among Displaced Persons Residing in a Temporary

Camp Following the August 1 999 Earthquake in Turkey

Dahna Batts-Obsome, Lina Balluz, and Mary Naughton (CDC) 401

Wind Storm-Related Disaster Deaths in The United States, 1994-1998

Julie Jacobson (CDC) 405

THEME 6. Natural Disasters in the World and the Lessons Learned

Urbanization and Urban Disaster Prevention

Tomomitsu FUJII (PWRI) 415

Effects of the August 17, 1999 Kocaeli (Izmit) Earthquake and the November 12, 1999

Duzce Earthquake, Turkey, on Dams

Ellis L. Krinitzsky and Mostafiz R. Chowdhury (USACE)

Ghassan Al-Chaar (Construction Engineering Research Laboratory) 427

Damage of Transportation Facility in the 1 999 Kocaeli and Duzce, Turkey Earthquakes

and the 1999 Chi-Chi, Taiwan Earthquake

Kazuhiko KAWASHIMA and Gaku SHOJI (Tokyo Institute of Technology) 439

The Performance of the Trans European Motorway Structures During the Nov. 12, 1999

Duzce Earthquake

Hamid Ghasemi and James D. Cooper (FHWA)

Roy Imbsen (IMBSEN & Associates, Inc.) 463

Building Damages in Taiwan Chi-Chi Earthquake

Masaomi TESHIGAWARA, Hiroshi FUKUYAMA, Hiroto KATO, and

Mazuo EMUKAI (BRI) 479

Damage of Earth Retaining Structures in the 1999 ChiChi, Taiwan Earthquake

Ken-ichi TOKIDA, Osamu MATSUO, and Shinya NAKAMURA (PWRI) 495

Hurricane Mitch Reconstruction Partnerships and Progress

William E. Roper (NIST) 513

Storm Surge Disaster by Typhoon No. 18 in 1999

Hiroyasu KAWAI (Port and Harbour Research Institute) 531

APPENDIX

Task Committee Reports 549

Technical Site Visits 575

4
AGENDA

5

AGENDA

32^° JOINT MEETING

PANEL ON WIND AND SEISMIC EFFECTS

16-19 May 2000

May 16 (Tuesday)

10:00 Opening Ceremonies

Call to order by John GROSS, Secretary-General, U.S. -Side Panel on Wind and Seismic

Effects

Welcome by James HILL, Deputy Director, Building and Fire Research Laboratory,

National Institute of Standards and Technology

Opening Rem.arks by Takao KURAMOCHI, Counselor for Science and Technology,

Embassy of Japan

Opening Remarks Donald FERGUSON, U.S. -side Coordinator, UJNR, International

Relations Advisor, U.S. Department of Agriculture

Remarks by S. Shyam SUNDER, Chairman U.S. -Side, Panel on Wind and Seismic

Effects and Chief, Structures Division, National Institute of Standards and Technology

Remarks by Tomomitsu FUJII, Chairman Japan-Side, Panel on Wind and Seismic Effects

and Director-General, PubHc Works Research Institute

Introduction of Japan Members by Japan Panel Chairman

Introduction of U.S. Members by U.S. Panel Chairman

Election of Joint Meeting Chairman

Adoption of Agenda

Adjourn

11:45 Group Photograph

12:00 Session 1 (Information Fechnology for Land Management and Natural Disaster

Prevention Countermeasures) Chairman - Mr. Tomomitsu Fujii

12:00 Web-Based GIS Support to Disaster Response; William ROPER, (NIST)

12:20 Draft Manual for Seismic Information System Development; Hideki SUGITA and

Tomofumi NOZAKI, (PWRI)

7
12:40 Lunch: Hosted by John Filson, Manager, Earthquake Hazards Program, U.S. Geological

Survey

13:40 Session 1 (Information Technology for Land Management and Natural Disaster

Prevention Countermeasures) - Continued

1 3 :40 The US Tsunami Mitigation Initiative, Michael BLACKFORD (NOAA)

14:00 Discussion

14:20 Break

14:40 Session 2 (Wind Engineering) Chairman - Mr. Tomornitsu Fujii

1 4:40 Design for Wind/Rain Response of Bridge Cables; Harold BOSCH (FHWA)

1 5:00 Study on Performance-Based Design Methods for Wind Effects on Bridges,

Hiroshi SATO and Ken-ichi OGI (PWRI)

15:20 Safe Rooms/Shelters; Cliff OLIVER (FEMA)

15:40 Discussion

16:00 Break

16:20 Task Committee Meetings

A: Strong Motion Data and Applications

B: Testing and Evaluation Procedures for Building Systems

C: High-Performance Structural Systems and Auto-Adaptive Media

D: Earthquake Engineering for Dams

E: Design for Wind and Wind Hazard Mitigation

F: Disaster Prevention Methods for Lifeline Systems

17:30 Adjourn for day

19:00 Banquet hosted by US-Side (National Academy of Sciences)

May 17 (Wednesday)

8:30 Session 3 (Special Topics: Natural Disasters in the World and the Lessons Learned, 1)

' Chairman - Dr. S. Shyam Sunder

8:30 Urbanization and Urban Disaster Prevention; Tomomitsu FUJII (PWRI)

8:50 Effects of the 17 August 1999 Kocaeli (Izmit) Earthquake and the 12 November

1 999 Duzce Earthquake, Turkey on Dams; Ellis KRINITSKY, Mostafiz

CHOWDHURY, and Ghassan AL-CHAAR (USAGE)

8
9:10 Damage of Transportation Facilities in the 1999 Kocaeli, Turkey Earthquake and

the Chi Chi, Taiwan Earthquake; Kaszuhiko KAWASHIMA and Gaku SHOJI

(Tokyo Institute of Technology)

9:30 The Performance of the Trans European Motorway Structures During the Nov.

12, 1999 Duzce Earthquake; Hamid GHASEMl, James Cooper (FHWA), Roy

IMBSEN (Imbsen & Associates, Inc.)

9:50 Building Damages in Taiwan Chi Chi Earthquake; Masaomi TESHIGAWARA, Hiroshi FUKUYAMA, Hiroto KATO, and Mizuo INUKAI (BRI)

10:10 Discussion

10:40 Break

11:00 Session 4 (Special Topics: Natural Disasters in the World and the Lessons Learned 2)

Chairman - Dr. S. Shyam Sunder

1 1 :00 Bridge Lessons in Taiwan - W. Phillip YEN (FHWA)

1 1 :20 Damage of Earth Retaining Structures During the 1999 Chi-Chi Earthquake; KenIchi TOKIDA, Osamu MATSUO, and Shinya NAKAMURA (PWRI)

1 1 :40 Information Infrastructure and Reconstruction After Hurricane Mitch; William

ROPER (NIST)

12:00 Storm Surge Disaster by Typhoon No. 18 in 1999; Hiroyasu KAWAI (PHRI)

(presented by Takashi NAGAO) 12:20 Discussion

12:50 Lunch: Hosted by Priscilla Nelson, Director, Division of Civil and Mechanical Systems,

National Science Foundation

13:50 Session 5 (Earthquake Engineering 1) Chairman - Dr. S. Shyam Sunder

13:50 Seismic Testing of a 1/20-Scale 2D Model of Koyna Dam; Robert HALL and

Mostafiz CHOWDHURY, Enrique MATHEU (USAGE)

14:10 Development of Smart Materials and Structural Systems - Research Plan on

Structural Systems, Sensing, and Effectors; Shunsuke OTANI (University of

Tokyo), Hisahiro HIRAISHI (BRI), Mitsumasa MIDORIKAWA (BRI), Masanori

IIBA (BRI), Masaomi TESHIGAWARA (BRI)

14:30 Nonlinear Analysis of Concrete Dams; Luis DEBEJAR, Robert HALL (USAGE)

14:50 Pre-Analytical Study on Damage Detection Test of Five-Story Steel Frame with

Simulated Damage; Masaomi TESHIGAWARA, Takashi KAMINOSONO,

Mizuo INUKAI, Hiroshi ISODA, Koichi MORITA, and Kazuya NOGUCHI

(BRI)

15:10 Limiting Pore Pressures in Deep Soil Deposits, M.E. HYNES and Richard

LEDBETTER (USAGE)

15:30 Soil Amplification Factor for Seismic Design of Buildings; Izuru OKAWA, Masanori IIBA, Mitsumasa MIDORIKAWA, Shin KOYAMA (BRI), and Kenji

MIURA (Kajima Corp.)

15:50 Discussion

9
16:20 Break

16:30 Task Committee Meetings

G: Seismic Information Systems

H: Soil Behavior and Stability during Earthquakes

I: Storm Surge and Tsunami

J: Wind and Earthquake Engineering Systems for Transportation Systems

17:30 Adjourn for Day

19:00 Individually hosted dinners

May 1 8 (Thursday)

8:30 Session 6 (Earthquake Engineering 2) Chairman - Mr. Tomomitsu Fujii

8:30 Probabilistic Analysis of Liquefaction of Soils; M. E. HYNES and Don YULE

(USACE)

8:50 Construction of a Base-Isolated House for Observation of Isolation Effects During

Earthquake and Wind; Yasuyuki YAMANOUCHI, Mitsum_asa MIDORIKAWA,

Masanori IIBA (BRI), Masayoshi IKENAGA (Giles Corp.), and Kenichi

MACHIDA (Sumitomo Forestry Co., Ltd.) presented by Izuru OHKAWA) 9:10 Structural Control; Michael RILEY and Fahim SADEK (NIST)

9:30 Seismic Design Ground Motion for Strait-Crossing Projects in Japan; Keiichi

TAMURA, Shojiro KATAOKA, Yoshihiro NAKAO (PWRI) and Masahiko

YASUDA (HSBA)

9:50 Discussion

10:20 Break

10:40 Session 7 (Earthquake Engineering 3) Chairman - Mr. Tomomitsu Fujii

10:40 Development of Seismic Isolation Technology for Underground Structures and

the Application; Shigeki UNJOH, Jun-Ichi HOSHIKUMA, Kazuhiro NAGAYA

(PWRI), Takeyasu SUZUKI (Kumagai-gumi Corp.)

1 1 :00 Earthquake Motions at an Embankment Dam Base and an Estimation Method of

Incident Seismic Waves Using the Observations; Yoshikazu YAMAGUCHI and

Tomoya IWASHITA (PWRI)

1 1 :20 Reproduction of Large-Scale IG Test on Dry Sand Deposit and Pile Foundation

using Centrifuge Modeling (Step-2); Tikahiro MINOWA (NRI), Masayoshi

SATO (Shimizu Co.), Yukiko SAITO (PWRI), Takaaki KAGAWA (Wayne State

University), and Akio ABE (Tokyo Soil Research Co.)

1 1 :40 Discussion

12:10 Break

10
12:20 Session 8 (Performance Based Design Methods 1) Chairman -Mr. Tomomitsu Fujii

12:20 Comparison of Performance-Based Design Codes for Bridges in Various

Countries; Michio OKAHARA and Masahiro NISHITANI (PWRI)

12:40 Recommended Changes to the AASHTO Specifications for the Seismic Design of

Bridges (NCHRP Project 12-49); Ian FRIEDLAND (ATC), Ronald MAYES

13:00 Lunch: Hosted by Jack Snell, Director, Building and Fire Research Laboratory, National

Institute of Standards and Technology

14:00 Session 8 (Performance Based Design Methods 1) - Continued

14:00 Performance-Based Design - The Big Picture; Tim SHECKLER, (FEMA)

14:20 Earthquake Resistant Design of Port Facilities; Takashi NAGAO and Atsushi

NOZU (PHRl)

14:40 CIS and Pre- and Post-Disaster Risk Assessment, Stuart NISHENKO (FEMA)

15:00 Discussion

15:30 Break

15:50 Session 9 (Performance-Based Design Methods 2) Chairman -Mr. Tomomitsu Fujii

15:50 Structural Housing Systems; Fahim SADEK (NIST)

16:10 Performance-Based Damage Assessment and Design Verification of U.S.

Housing in Extreme Earthquake and Hurricane Events; Jay CRANDALL (NAHB

Research Center) and William FREEBORNE (HUD)

16:30 American Lifeline Alliance and Performance-Based Guidelines for Lifelines and

Utilities; Woody SAVAGE (FEMA)

16:50 Discussion

17:10 Adjourn

18:15 US-Side Hosted Dinner at Smokey Glen

May 19 (Fri)

8:30 Session 10 (Public Health) Chairman - Dr. S. Shyam Sunder

8:30 A Proposed Model for Multidisciplinary Needs Assessment in Post Disaster

Settings; Josephine MALILAY (CDC)

8:50 Living Difficulties After Earthquakes and the Related Phenomena; Keishi

SHIONO (Nagaoka College) (presented by Keiichi TAMURA) 9:10 Predictors for People's Response to Tornado Warning, Arkansas, March 1, 1997;

Lina BALLUZ (CDC)

11
9:30 Post-Disaster Carbon Monoxide Poisoning; An Emerging Health Problem;

Randolph DALEY (CDC)

9:50 Assessment of Public Health Needs Among Displaced Persons Residing in a Temporary Camp Following the August 1999 Earthquakes in Turkey; Dahna

BATTS-QSBORNE (CDC)

10:10 Wind-Related Climactic Disaster-Related Deaths in the United States, 1994-1998;

Julie JACOBSEN (CDC)

10:30 Discussion

10:50 Break

11;05 Workshop Reports

12:00 Reports of Task Committees 1

13:00 Lunch: Hosted by Craig Wingo, Director of the National Earthquake Program, Federal

Emergency Management Agency

14:00 Reports of Task Committees 2

15:00 Adoption of Final Resolutions

15:50 Break

16:00 Closing Ceremonies

Call to Order by John GROSS, Secretary-General, US-side Panel

Closing Remarks by Tomomitsu FUJII, Chairman Japan-side Panel

Closing Remarks by S. Shyam SUNDER, Chairman US-side Panel

16:30 Conclusion of 32"** Joint Panel Technical Sessions

19:00 Japan-Side Hosted Dinner at Japan Inn

12
Technical Site Visits

32"^ Joint Panel on Wind and Seismic Effects

Saturday, 20 May Depart Washington Dulles at 9:20 am on United Airlines Flight

UA1919 Arrive San Juan, Puerto Rico 1:11 pm (local time)

Sightseeing in Old San Juan

Stay at Marriott Condado Beach Hotel

Sunday, 21 May Visit Arecibo Observatory Radio Telescope (NSF)

Cultural visit to Rio Camuy Cave Park

Travel to Ponce, Puerto Rico

Stay at Holiday Inn in Ponce

Monday, 22 May Visit Cerrillos and Portugues dam projects (USACE)

Cultural tour of Serralles Castle and Parque de Bombas

Return to San Juan

Stay at Caribe Hilton in San Juan

Tuesday, 23 May Visit Tren Urbano (mass transit system - under construction)

Japan-side delegation departs from San Juan

13

!

I

List of Panel Members

15

JAPAN-SIDE PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

STEERING COMMITTEE MEMBERS

Mr. Tomomitsu Fujii

Chairman, Japan-side Panel on Wind and Seismic

Effects Director-General, Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2211

Fax: 0298-64-2148

E-mail: fujii310@pwri.go.jp

Dr. Michio Okahara

Secretary-General, Japan-side Panel on Wind and

Seismic Effects

Director, Structure and Bridge Department

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-221

1

Fax: 0298-64-0565

E-mail: okahara@pwri.go.jp

Mr. Hiroshi Akiyama

Director, Geographic Department

Geographical Survey Institute Ministry of Construction

1, Kitasato, Tsukuba-shi, Ibaraki-ken 305-081

1

Tel:0298-64-llll

Fax:0298-64-1804

Mr. Yoshitoki Hashimoto

Director, Disaster Countermeasure Office

Coast Administration Disaster Prevention Division

Ports and Harbors Bureau

Ministry of Transport

2-1-3, Kasumigaseki, Chiyoda-ku, Tokyo 100-0013

Tel: 03-3580-7021

Fax: 03-5511-8280

Dr. Hisahiro Hiraishi

Director, Codes and Evaluation Research Center

Building Research Institute Ministry of Construction

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6633

Fax: 0298-64-6773

E-mail: hiraishi@kenken.go.jp

Dr. Susumu lai Director, Earthquake Disaster Prevention Laboratory

Port and Harbour Research Institute Ministry of Transport

3-1-1, Nagase, Yokosuka-shi, Kanagawa-ken 239-

0826

Tel: 0468-44-5030

Fax: 0468-44-0839

E-mail: iai@ipc.phri.go.jp

Dr. Toshio Iwasaki

President, Civil Engineering Research Laboratory

1- 18 Kandasuda-cho

Chiyoda-ku, Tokyo 101-0041

Tel: 03-3254-9481

Fax: 03-3254-9448

Mr. Kouichi Iwase

Office for Disaster Prevention Research

Research and Development Bureau

Science and Technology Agency

Prime Minister's Office

2-2-1, Kasumigaseki, Chiyoda-ku, Tokyo 100-0013

Tel: 03-3503-8164

Fax: 03-3503-8169

Dr. Takashi Kaminosono

Associate Director for International Codes and

Standards

Codes and Evaluation Research Center

Building Research Institute Ministry of Construction

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel:0298-64-6665

Fax:0298-64-6774

E-mail: kamino@kenken.go.jp

Dr. Kazumasa Kato

Director, Hydraulic Engineering Division

Port and Harbour Research Institute Ministry of Transport

3- 1-1, Nagase, Yokosuka-shi, Kanagawa-ken 239-

0826

Tel: 0468-44-5009

Fax:0468-41-3888

17
Mr. Shirou Kato

Director, Road Disaster Prevention Section

Road Bureau

Ministry of Construction

2-1-3, Kasumigaseki, Chiyoda-ku, Tokyo 100-0013

Tel: 03-3580-431

1

Fax:03-5251-1949

E-mail: katouOzv@road.hs.moc.go.jp

Prof. Kazuhiko Kawashima

Professor, Department of Civil Engineering

Tokyo Institute of Technology

2- 12-1, O-okayama, Meguro-ku, Tokyo 152-0033

Tel: 03-5734-2922

Fax: 03-3729-0728

E-mail: kawasima@cv.titech.ac.jp

Mr. Osamu Matsuo

Head, Soil Dynamics Division

Earthquake Disaster Prevention Research Center

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2933

Fax: 0298-64-2576

E-mail: matsuo@pwri.go.jp

Mr. Chikahiro Minowa

Cooperative Research Officer

National Research Institute for Earth Science and

Disaster Prevention

3- 1, Tennodai, Tsukuba-shi, Ibaraki-ken 305-0006

Tel: 0298-51-1611

Fax: 0298-51-5658

E-mail: minowa@geo.bosai.gov.jp

Mr. Toshitaka Miyata

Director, International Affairs Division

Economic Affairs Bureau

Ministry of Construction

2- 1-3, Kasumigaseki, Chiyoda-ku, Tokyo 100-0013

Tel: 03-3580-43 11 Fax:03-3502-3955

E-mail: miyat01e@econ.hs.moc.go.jp

Dr. Takashi Nagao

Chief of Geotechnical Earthquake Engineering

Laboratory

Port and Harbour Research Institute Ministry of Transport

3- 1-1, Nagase, Yokosuka-shi, Kanagawa-ken 239-

0826

Tel: 0468-44-5028

Fax: 0468-44-0839

E-mail: nagao@ipc.phri.go.jp

Dr. Nobuyuki Narita

President, Japan Steel Structure Corporation

3-3-1, Marunouthi, Chiyoda-ku, Tokyo 100-0005

Tel: 03-3212-0875

Fax: 03-3212-0878

Mr. Kazuhiro Nishikawa

Head, Bridge Division

Structure and Bridge Department

Public Works Research Institute Ministry of Construction

I , Asahii Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2905

Fax: 0298-64-0565

E-mail: knishikawa@pwri.go.jp

Dr. Isao Nishiyama

Head, Housing Construction Division

Production Department

Building Research Institute Ministry of Construction

1, Tatehara, Tsukuba-shi, ibaraki-ken 305-0802

Tel: 0298-64-2151

Fax: 0298-64-6774

E-mail: isao@kenken.go.jp

Mr. Tomiyoshi Ogawa

Director, Building Disaster Prevention Section

Housing Bureau

Ministry of Construction

2- 1-3, Kasumigaseki, Chiyoda-ku, Tokyo 100-0013

Tel: 03-3580-4311

Fax: 03-3580-7050

E-mail: ogawa0a2@house.hs.moc.go.jp

Mr. Keiichi Ohtani

Director, Disaster Prevention Research Division

National Research Institute for Earth Science and

Disaster Prevention

Science and Technology Agency

3- 1, Tennodai, Tsukuba-shi, Ibaraki-ken 305-0006

Tel: 0298-51-1611 ex.321

Fax: 0298-52-8512

E-mail: ohtani@knetgk.k-net.bosai.go.jp

Dr. Hisashi Okada

Director, Structural Engineering Department

Building Research Institute Ministry of Construction

1 , Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6641

Fax: 0298-64-6773

E-mail: okada@kenken.go.jp

18
Prof. Tsuneo Okada

Professor, Department of Architecture and Building

Engineering

Shibaura Institute of Technology

3-9-14 Shibaura, Minato-ku, Tokyo 108-0023

Tel: 03-5476-2452

Fax: 03-5476-2446

Dr. Izuru Ohkawa

Head, Construction Techniques Division

Production Department

Building Research Institute Ministry of Construction

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-2151

Fax: 0298-64-6777

E-mail: okawa@kenken.go.jp

Mr. Kazuo Okayama

Director, Earthquake Disaster Countermeasure

Division

Disaster Prevention Bureau

National Land Agency

1-2-2, Kasumigaseki, Chiyoda-ku, Tokyo 100-0013

Tel: 03-3593-33 11 Fax: 03-3501-5199

Dr. Hiroshi Sato Head, Structure Division

Structure and Bridge Department

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2874

Fax: 0298-64-0565

E-mail: hsato@pwri.go.jp

Dr. Takahiro Sugano

Chief, Structural Dynamics Laboratory

Structural Engineering Division

Port and Harbour Research Institute Ministry of Transport

3-1-1, Nagase, Yokosuka-shi, Kanagawa-ken 239-

0826

Tel: 0468-44-5029

Fax: 0468-44-0839

E-mail: macsuga@ipc.phri.go.jp

Dr. Hideki Sugita

Head, Earthquake Disaster Prevention

Technology Division

Earthquake Disaster Prevention Research Center

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-3245

Fax: 0298-64-0598

E-mail: sugita@pwri.go.jp

19

Mr. Hidetaka Takagi

Director, Structure Department

Civil Engineering Research institute Hokkaido Development Bureau

Prime Minister's Office

1-3, Hiragishi, Toyohira-ku, Sapporo-shi

Hokkaido 062-093

1

Tel: 01 1-841-5175

Fax: 01 1-820-2714

Mr. Ken-ichi Takahashi

International Cooperation Coordinator, International

Affairs Division,

Science and Technology Promotion Bureau

Science and Technology Agency

2-2-1, Kasumigaseki, Chiyoda-ku, Tokyo 100-0013

Tel: 03-3593-0362

Fax: 03-3501-5909

E-mail: katakaha@sta.go.jp

Mr. Syuji Takasu

Director, Dam Department

Public Works Research institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2832

Fax: 0298-64-0164

E-mail: takasu@pwri.go.jp

Dr. Keiichi Tamura

Head, Ground Vibration Division

Earthquake Disaster Prevention Research Center

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2926

Fax: 0298-64-0598

E-mail: tamura@pwri.go.jp

Dr. Masaomi Teshigawara

Head, Structure Division

Department of Structural Engineering

Building Research Institute Ministry of Construction

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6753

Fax: 0298-64-6773

E-mail: teshi@kenken.go.jp

Mr. Ken-ichi Tokida

Director, Earthquake Disaster Prevention Research

Center

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2829

Fax: 0298-64-0598

E-mail: tokida@pwri.go.jp
Mr. Ken-ichi Torii

Head, Coast Division

River Department

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2211

Fax:0298-64-1168

E-mail: torii@pwri.go.jp

Dr. Hajime Tsuchida

Executive Counselor, Nippon Steel Co., Ltd. 2-6-3, Ohte-machi, Chiyoda-ku, Tokyo 100-0004

Tel: 03-3275-5894

Fax: 03-3275-5648

Dr. Takaaki Uda

Director, River Department

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2211

E-mail: uda@pwri.go.jp

Dr. Shigeki Unjoh

Head, Earthquake Engineering Division

Earthquake Disaster Prevention Research Center

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-4966

Fax: 0298-64-4424

E-mail: unjoh@pwri.go.jp

Mr. Shoin Yagi

Director, Typhoon Research Department

Meteorological Research Institute Japan Meteorological Agency

1-1, Nagamine, Tsukuba-shi, Ibaraki-ken 305-0052

Tel: 0298-53-8663

Fax: 0298-53-8735

Dr. Yoshikazu Yamaguchi

Head, Fill-type Dam Division

Dam Department

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2413

Fax:0298-64-0164

E-mail: yamaguti@pwri.go.jp

Dr. Hiroyuki Yamanouchi

Director, Codes and Evaluation Research Center

Building Research Institute Ministry of Construction

1, Takehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6688

Fax: 0298-64-67701

E-mail: yamanoch@kenken.go.jp

Mr. Takashi Yokota

Senior Assistant for Disaster Prevention

Planning Division

Administration Department

Japan Meteorological Agency

1-3-4, Ohte-machi, Chiyoda-ku, Tokyo 100-0013

Tel: 03-3212-8341

Fax:03-3212-2453

Dr. Akio Yoshida

Director, Seismology and Volcanology

Research Department

Meteorological Research Institute Japan Meteorological Agency

1- 1, Nagamine, Tsukuba-shi, Ibaraki-ken 305-0052

Tel: 0298-53-8675

Fax: 0298-53-8735

Dr. Kiyofumi Yoshino

Director, Disaster Management Division

River Bureau

Ministry of Construction

2- 1-3, Kasumigaseki, Chiyoda-ku, Tokyo 100-0013

Tel: 03-3580-4311 Ext. 3431

Fax: 03-5251-1946

E-mail: yanagOan@river.hs.moc.go.jp

20
SECRETARIES COMMITTEE MEMBERS

Dr. Michio Okahara

Secretary-General, Japan-side Panel on Wind and

Seismic Effects

Director, Structure and Bridge Department

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-221

1

Fax: 0298-64-0565

E-mail: okahara@pwri.go.jp

Mr. Jiro Fukui

Head, Foundation Engineering Division

Structure and Bridge Department

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2211

Fax: 0298-64-0565

E-mail: flikui@pwri.go.jp

Dr. Hisahiro Hiraishi

Director, Structural Engineering Department

Building Research Institute Ministry of Construction

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6633

Fax: 0298-64-6773

E-mail: hiraishi@kenken.go.jp

Dr. Masanori liba Head, Aerodynamics Division

Structural Engineering Department

Building Research Institute Ministry of Construction

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6654

Fax: 0298-64-6773

E-mail: iiba@kenken.go.jp

Mr. Kenji Ikeda

Director, Structures Division

Structures Department.

Civil Engineering Research Institute Hokkaido Development Bureau

Prime Minister's Office

1-3, Hiragishi, Toyohira-ku, Sapporo-shi

Hokkaido 062-0931

Tel: 011-841-11 11 Fax:011-820-2714

E-mail: ikedak@ceri.go.jp

Mr. Jun-.ich Kaneko

Head, Planning Division

Geographic Department

Geographical Survey Institute Ministry of Construction

1, Kitasato, Tsukuba-shi, Ibaraki-ken 305-081

1

Tel:0298-64-llll

Fax:0298-64-1804

Mr. Osamu Matsuo

Head, Soil Dynamics Division

Earthquake Disaster Prevention Research Center

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2933

Fax: 0298-64-2576

E-mail: matsuo@pwri.go.jp

Mr. Chikahiro Minowa

Cooperative Research Officer

National Research Institute for Earth Science and

Disaster Prevention

3-1, Tennodai, Tsukuba-shi, Ibaraki-ken 305-0006

Tel: 0298-51-1611

Fax: 0298-51-5658

E-mail: minowa@geo.bosai.gov.jp

Dr. Takashi Nagao

Chief of Geotechnical Earthquake Engineering

Laboratory

Port and Harbour Research Institute Ministry of Transport

3-1-1, Nagase, Yokosuka-shi, Kanagawa-ken 239-

0826

Tel: 0468-44-5028

Fax: 0468-44-0839

E-mail: nagao@ipc.phri.go.jp

Mr. Ryouji Namikawa

Head, International Cooperation Division

Planning and Research Administration Department

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-4412

Fax: 0298-64-4322

E-mail: namikawa@pwri.go.jp

21
Mr. Kazuhiro Nishikawa

Head, Bridge Division

Structure and Bridge Department

Public Works Research Institute Ministry of Construction

1 , Asaiii, Tsukuba-slii, Ibaraki-ken 305-0804

Tel: 0298-64-2905

Fax: 0298-64-0565

E-mail: knishika@pwri.go.jp

Dr. Nobuyuki Ogawa

Head, Earthquake Engineering Laboratory

National Research Institute for Earth Science and

Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Tsukuba-shi, Ibaraki-ken 305-0006

Tel: 0298-51-1611

Fax:0298-51-5658

E-mail: ogawa@geo.bosai.go.jp i;

Mr. Keiichi Ohtani

Director, Disaster Prevention Research Division

National Research Institute for Earth Science and

Disaster Prevention

Science and Technology Agency ' •

3-1, Tennodai, Tsukuba-shi, Ibaraki-ken, 305-0006

Tel: 0298-51-1611 Ext. 321

Fax:0298-52-8512

E-mail: ohtani@knetgk.k-net.bosai.go.jp

Dr. Hisashi Okada

Director, Structural Engineering Department

Building Research Institute Ministry of Construction

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6641

Fax:0298-64-6773 ' •

; E-mail: okada@kenken.go.jp

Dr. Hiroshi Sato Head, Structure Division

Structure and Bridge Department

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2874

Fax: 0298-64-0565

E-mail: hsato@pwri.go.jp

Mr. Masaya Sato

Director, Research Planning and Coordination

Section, Administration Division

Civil Engineering Research Institute Hokkaido Development Bureau

Prime Minister's Office

1-3, Hiragishi, Toyohira-ku, Sapporo-shi

Hokkaido 062-093

1

Tel:011-841-llll

Fax:011-820-2714

Dr. Takahiro Sugano

Chief, Structural Dynamics Laboratory

Structural Engineering Division

Port and Harbour Research Institute Ministry of Transport

3-1-1, Nagase, Yokosuka-shi, Kanagawa-ken 239-

0826

Tel: 0468-44-5029

Fax: 0468-44-0839

E-mail: macsuga@ipc.phri.go.jp

Dr. Hideki Sugita

Head, Earthquake Disaster Prevention Technology

Division

Earthquake Disaster Prevention Research Center

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-3245

Fax: 0298-64-0598

E-mail: sugita@pwri.go.jp

Dr. Keiichi Tamura

Head, Ground Vibration Division

Earthquake Disaster Prevention Research Center

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2926

Fax: 0298-64-0598

E-mail: tamura@pwri.go.jp

Mr. Ken-ichi Tokida

Director, Earthquake Disaster Prevention Research

Center

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2829

Fax: 0298-64-0598

E-mail: tokida@pwri.go.jp

22
Mr. Ken-ichi Torii

Head, Coast Division

River Department

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2327

Fax: 0298-64-1168

E-mail: sato@pwri.go.jp

Dr. Shigeki Unjoh

Head, Earthquake Engineering Division

Earthquake Disaster Prevention Research Center

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-4966

Fax: 0298-64-4424

E-mail: unjoh@pwri.go.jp

Dr. Yoshikazu Yamaguchi

Head, Fill-type Dam Division

Dam Department

Public Works Research Institute Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2413

Fax: 0298-64-0164

E-mail: yamaguti@pwri.go.jp

Mr. Toyotaro Yamauchi

Head, The First Research Laboratory

Seismology and Volcanology Research Department

Meteorological Research Institute Japan Meteorological Agency

Ministry of Transport

1-1, Nagamine, Tsukuba-shi, Ibaraki-ken 305-0052

Tel: 0298-51-3730

Fax: 0298-53-8735

ASSOCIATE MEMBERS

Dr. Minoru Fujiwara

President, Strait Crossing Highway Research

Association

1-4-13, kyobashi, chuo-ku,Tokyo 104-0031

Tel: 03-3245-9840

Fax: 03-3245-9844

Dr. Masami Fukuoka

President, Management System Assessment Center,

2- 17-7, Kyoubashi, Chuo-ku, Tokyo 104-0031

Tel:03-3535-4021

Fax: 03-3535-4024

Dr. Masaya Hirosawa

Professor, Kogakuin University

1-24-2-91, Nishi-shinjuku, Shinjuku-ku, Tokyo 163-

0023

Tel: 03-3342-1211

Fax:03-3340-0149

Dr. Shiro Ibukiyama

President, Kogyokusha College of Technology

5-14-2, Nishigotanda, Shinagawa-ku, Tokyo 141- 0031

Tel: 03-3493-5671

Fax: 03-3495-4071

Dr. Kaoru Ichihara

Adviser, Central Consul Tant Inc. 2- 16-2, Minami-Kamata, Ohta-ku, Tokyo 144-0035

Tel:03-5703-6161

Fax:03-5703-6151

Dr. Ryuichi lida President, Japan Dam Engineering Center

2-4-5, Azabudai, Minato-ku, Tokyo 106-0041

Tel: 03-3433-7811

Fax: 03-3432-6204

Dr. Takashi lijima

President, Sekisui-jushi Co., Ltd.

1- 11-1, Kaigan, Minato-ku, Tokyo 105-0022

Tel: 03-5400-1801

Fax: 03-5400-1805

Mr. Yasutake Inoue

President, Fukuoka-Kitakyusyu Expressway Public

Cooperation

2-7-53, Higashihama, Higashi-ku, Fukuoka-shi,

Fukuoka-ken 812-0055

Tel: 092-631-3281

Dr. Toshio Iwasaki

President, Civil Engineering Research Laboratory

1-18 Kandasuda-cho,Chiyoda-ku, Tokyo

101-0041

Tel: 03-3254-9481

Fax: 03-3254-9448

Mr. Shunichiro Kamijo

President, Mitsui Construction Corporation,

1-2-3, Ohte-machi, Chiyoda-ku,

Tokyo 1 00-0004

Tel: 03-5223-3915

Fax: 03-5223-3920

23

/
Mr. Kenji Kawakami

Adviser, Kubota Corporation

3-1-3, Nihonbashi-muromachi, Cyuoh-i<u, Tokyo

103-0022

Tel: 03-3245-3660

Fax: 03-3245-3454

Prof. Eiichi Kuribayashi

10-2-801, Nishihouwa, Yayoi-cho, Toyohashi-shi

Aichi-ken 441-8122 > Tel: 0532-47-0429

Dr. Tatsuro Murota

President, General Building Research Corporation

of Japan

5-8-1, Fujishirodai, Suita-shi, Osaka-fu 565-0873

Tel: 06-6872-0391

Fax: 06-6872-0784

Mr. Mitsuru Nagao

President, Japan Construction Mechanization

Association

3-5-8, Shibakoen, Minato-ku, Tokyo 105-001

1

Tel: 03-3433-1501

Fax: 03-3432-0289

Prof Kiyoshi Nakano

979-66, Kumada, Minaminasu-machi, Nasu-gun,

Tochigi-ken

Tel: 0287-88-0022

Dr. Kazuto Nakazawa

Adviser, Public Works Research Center

1-6-4, Taito, Taito-ku, Tokyo 1 10-0016

Tel: 03-3835-3609

Fax: 03-3275-5648

Dr. Nobuyuki Narita

President, Japan Steel Structure Corporation

3-3-1, Marunouthi, Chiyoda-ku,

Tokyo 100-0005

Tel: 03-3212-0875

Fax: 03-3212-0878

Dr. Setuo Noda

President, Coastal Development Institute of

Technology

3-16, Hayabusa-cho, Chiyoda-ku, Tokyo

102-0092

Tel: 03-3234-5861

Fax: 03-3234-5877

Dr. Shin Okamoto

Central Research Institute for Construction

Technology

Japan Association of Representative

General Contractors

1-6-34, Kounan, Minato-ku, Tokyo 108-0075

Tel: 03-3458-1011

Fax: 03-3458-6321

Mr. Yoshijiro Sakagami

Receive, Japan Development Construction

Corporation,

4-9-9, Akasaka, Minato-ku, Tokyo 107-0052

Tel: 03-5410-5730

Fax: 03-3403-9039

Dr. Tadahiko Sakamoto

President, Japan Dam Engineering Center

2-4-5, Azabudai, Minato-ku, Tokyo 106-0041

Tel: 03-3433-7811

Fax: 03-3432-6204

Prof Yasushi Sasaki

Professor, Department of Civil and Environmental

Engineering

Hiroshima University

1- 4-1, Kagamiyama, Higashihiroshima-shi,

Hiroshima-ken 739-0046

Tel: 0824-24-7783

Fax: 0824-24-7783

E-mail: ysasaki@ue.ipc.hiroshima-u.ac.jp

Dr. Yukihiko Sumiyoshi

Executive Counselor, Nippon Steel Co., Ltd.

2-6-3, Ohte-machi, Chiyoda-ku,

Tokyo 100-0004

Tel: 03-3275-5868

Fax:03-3275-5636

Mr. Jiro Taguchi

President, Construction Research Institure

1 1-8, Ohdenma-chou, Nihonbashi, Chuo-ku,

Tokyo 103-0011

Tel: 03-3663-2411

Fax: 03-3633-2417

Dr. Hiroshi Takahashi

4-22-1, Nishihara-machi, Fuchuu-shi, Tokyo 183-

0046

Tel: 0425-72-3227

Dr. Masateru Tominaga

President, Public Works Research Center

1-6-4, Taito, Taito-ku, Tokyo 1 10-0016

Tel: 03-3835-3609

Fax: 03-3275-5648

24
Dr. Hajime Tsuchida

Executive Counselor, Nippon Steel Co., Ltd. 2-6-3, Ohte-machi, Chiyoda-ku, Tokyo 100-0004

Tel: 03-3275-5894

Fax: 03-3275-5648

Mr. Seize Tsuji

President, Japan Highway Public Corporation

3-3-2, Kasumigaseki, Chiyoda-ku, Tokyo 100-8979

Tel: 03-3506-0123

Fax: 03-3506-0396

Dr. Makoto Watabe

2-25-5, Tokiwadai, Itabashi-ku, Tokyo 174-0071

Tel: 03-3960-0374

Prof. Kouichi Yokoyama

Professor, Urban and Civil Engineering

Faculty of Engineering, Ibaraki University

4- 12-1, Nakanarisawa-machi, Hitachi-shi,

Ibaraki-ken 316-8511

Tel: 0294-38-5161

Fax: 0294-35-8146

25
UNITED STATES-SIDE PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

Dr. S. Shyam Sunder

Chairman, U.S. -side Panel

Chief, Structures Division

Building and Fire Research Laboratory

National Institute of Standards and

Technology

100 Bureau Drive, Stop 86 11 Gaithersburg, MD 20899-861

1

301-975-6061; Fax 301-869-6275

E-mail: sunder@nist.gov

Dr. John L. Gross

Secretary-General, U.S. -side Panel

Leader, Structural Systems and Design

Group

Building and Fire Research Laboratory

National Institute of Standards and

Technology

100 Bureau Drive, Stop 86 11

Gaithersburg, MD 20899-8611

301-975-6068; Fax 301-869-6275

E-mail: john.gross@nist.gov

Mr. Stephen A. Cauffman

Research Engineer

Structural Systems and Design Group

Building and Fire Research Laboratory

National Institute of Standards and

Technology

100 Bureau Drive, Stop 861

1

Gaithersburg, MD 20899-8611

301-975-6051; Fax 301-869-6275

E-mail: stephen.cauffman@nist.gov

Dr. Daniel P. Abrams

Director, Mid-America Earthquake Center

Hanson Engineers Professor of Civil

Engineering

University of Illinois

1 24 1 Newmark Laboratory

205 North Matthews Avenue

Urbana, IL 61801

217-244-6302; Fax 217-333-3821

E-mail: d-abrams@uiuc.edu

Dr. Kharaiti L. Abrol

Principal Structural Engineering Consultant

Department of Veterans Affairs

810 Vermont Avenue, NW, Room 475

Washington, DC 20420

202-565-5579; Fax 202-565-5478

E-mail: abrkha@hq.med.va.gov

Dr. John Ake

Geophysicist

Seismotechtonics and Geophysics Section

Bureau of Reclamation

U.S. Department of Interior

Code D-8330

P.O. Box 25007

Denver, CO 80225

303-236-4195 x276; Fax 303-236-9127

E-mail: jake@seismo.usbr.gov

Mr. John Baals

Seismic Safety Coordinator

P.O. Box 25007 (D-8110)

Bureau of Reclamation

U.S. Department of Interior

Denver, CO 80225

303-236-3999 x534; 303-236-9099

E-mail: ibaals@do.usbr.gov

26
Dr. Celso S. Barrientos

Supervisory Physical Scientist

National Environmental Satellite Data

Information Service - Code E/RA28

National Oceanic and Atmospheric

Administration

U.S. Department of Commerce

5200 Auth Road

Camp Springs, MD 20746

301-763-8102; Fax 301-763-8020

E-mail: cbarrientos@nesdis.noaa.gov

Dr. Eddie N. Bernard

Director, Pacific Marine Environmental

Laboratory

National Oceanic and Atmospheric

Administration

U.S. Department of Commerce

7600 Sand Point Way, NE

BINC15700/Building3

Seattle, WA 98115-0070

206-526-6800; Fax 206-526-6815

Email: bemard@pmel.noaa.gov

Mr. Mike Blackford

lOC-NWS/INTERNATIONAL

Director, International Tsunami Information

Center

737 Bishop Street, Suite 2200

Honolulu, HI 96813-3213

808-532-6423; Fax 808-532-5576

E-mail: Michael.Blackford@noaa.gov

Dr. David M. Boore

U.S. Geological Survey

345 Middlefield Road, MS 977

Menlo Park, CA 94025

650-329-5616; Fax 650-329-5163

E-mail: boore@samoa.wr.usgs.gov

Dr. Roger D. Borcherdt

Branch of Seismology

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

650-329-5619; Fax 650-329-5163

E-mail: borcherdt@samoa.wr.usgs.gov

Dr. Mehmet K. (^elebi

Research Civil Engineer

Branch of Earthquake and Geomagnetic

Information

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

650-329-5623; Fax 650-329-5163

E-mail: celebi@usgs.gov

Mr. Harish Chander

U. S. Department of Energy (EH-31)

CXXI, Room 2016(GTN)

1 990 1 Germantown Road

Germantown, MD 20874-1290

301-903-6681; Fax 301-903-8693

E-mail: harish.chander@hq.doe.gov

Dr. Ken P.Chong

Program Director

Civil and Mechanical Systems Division

4201 Wilson Boulevard, Room 545

Arlington, VA 22230

703-292-7008; Fax 703-292-9053

E-mail: kchong@nsf.gov

Mr. James D. Cooper

Chief, Structures Division, FINR-10

Federal Highway Administration

U.S. Department of Transportation

6300 Georgetown Pike

McLean, VA 22101

202-493-3023; Fax 202-493-3442

E-mail: iim.cooper@fhwa.dot.gov

27
Mr. William Freebome

Department of Housing and Urban

Development

Room 8132

451 Seventh Street, SW

Washington, DC 20410

202-708-4370 x5725; Fax 202-708-5873

E-mail: william_e. freeborne@hud.gov

Mr. Peter E. Gurvin

Director, Building Design and Engineering

Division

Foreign Building Operations

Building SA-6, Room 335

U.S. Department of State

Washington, DC 20520

703-875-61 17; Fax 703-875-6204

E-mail: peter.e.gurvin@dos.us-state.gov

Dr. Robert L. Hall

Chief, Structural Analysis Group

U.S. Army Engineer Research and

Development Center

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-2567; Fax 601-634-3412

E-mail: hallr3 @mai1 . v^^es . army. mi1

Dr. Allen M. Hittelman

Chief, Solid Earth Geophysics Division

E/GC 1, National Geophysical Data Center

NOAA, NESDIS

325 Broadway

Boulder, CO 80303-3328

303-497-6591; 303-497-6513

E-mail: allen.m.hittelman@noaa.gov

Mr. Larry C. Hultengren

Senior Structural Engineer

Office of Civil/Structural Engineering

Foreign Building Operations

U.S. Department of State

Building SA-6, Room 346

Washington, DC 20520

703-875-6194; Fax 703-875-6204

E-mail: hultengremlc@state.gov

Dr. Mary Ellen Hynes

Chief, Earthquake Engineering and

Geophysics Branch

Geotechnical Laboratory

U.S. Army Engineer Research and

Development Center

3909 Halls Ferry Road

Vicksburg, MS 39180

601-634-2280; 601-634-3453

E-mail: hynesm@exl . wes.army.mil

Dr. William B. Joyner

Geophysicist

Branch of Seismology

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

MenloPark, CA 94025

650-329-5640; Fax 650-329-5163

E-mail: joyner@samoa.wr.usgs.gov

Dr. George Lee

Director, Multidisciplinary Center for

Earthquake Engineering Research (MCEER)

State University of New York at Buffalo

Red Jacket Quadrangle

Buffalo, NY 14261

716-645-3391; Fax 716-645-3399

E-mail: gclee@acsu.buffalo.edu

Dr. H. S. Lew

Structures Division

Building Fire and Research Laboratory

National Institute of Standards and

Technology

100 Bureau Drive, Stop 861

1

Gaithersburg, MD 20899-861

1

301-975-6060; Fax 301-869-6275

E-mail: hai.lew@nist.gov

28
Dr. Shih-Chi Liu

Program Director

Civil and Mechanical Systems Division

National Science Foundation

4201 Wilson Boulevard, Room 545

Arlington, VA 22230

703-292-8360; Fax 703-292-9053

E-mail: sliu@nsf.gov

Dr. Josephine Malilay

Disaster Assessment and Epidemiology

Section

National Center for Environmental Health

1600 Clifton Road, NE (Mailstop E-23)

Atlanta, GA 30333

404-639-2589; Fax 404-639-2565

E-mail: jym7@cdc.gov

Dr. Francis G. McLean

Value Engineering Program Manager

Code D-8170

P.O. Box 25007

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303-445-3091; Fax 303-445-6475

E-mail: FMCLEAN@do.usbr.gov

Dr. Raymond E. Meyer

Department of State

SA14/Room 1705

USAID

Washington, DC 20523-1443

202-712-1078; Fax 202-216-3707

E-mail: Rmeyer(@usaid.gov

Dr. Jack Moehle

Director, Pacific Earthquake Engineering

Research Center (PEER)

1301 South 46''' Street Richmond CA 94804-4698

510-231-9554; 510-231-9471

E-mail: moehle@eerc.berkeley.edu

Mr. Ugo Morelli

Policy Manager

Federal Emergency Management Agency

500 C Street, SW

Washington, DC 20472

202-646-2810; Fax 202-646-2577

E-mail: ugo.morelli@fema.gov

Mr. Howard D. Nickerson

Earthquake Engineering and Weapons

Specialist

Naval Facilities Engineering Command

Washington Navy Yard

901 M Street, SE

Code 00CE9

Washington, DC 203745063

Phone: 202-443-8759

Fax: 202-433-8777

E-mail: hdnickerson@nfesc.navy.mil

Dr. Stuart Nishenko

Federal Emergency Management Agency

500 C Street NW

Washington, DC 20472

202-646-3945; Fax 202-646-4596

E-mail: stuart.nishenko@nist.gov

Mr. Tom Post

Chief, Earthquake Engineering Division

California Department of Transportation

(CALTRANS)

P.O. Box 942873

Sacramento, CA 95814

916-227-8728; Fax 916-227-8898

E-mail: tpost@trmx3 .dot. ca.gov

Dr. Phillip Yen

Federal Highway Administration

U.S. Department of Transportation

6300 Georgetown Pike

McLean, VA 22101

202-493-3056; Fax 202-493-3442

E-mail: wen-huei.yen@fhwa.dot.gov

29
ALTERNATE MEMBERS

Dr. Clifford Astill

Program Director

Civil and Mechanical Systems Division

National Science Foundation

4201 Wilson Boulevard, Room 545

Arlington, VA 22230

703-292-8360; Fax 703-292-9053

E-mail: castill@nsf.gov

Mr. Michael Changery

Chief, Global Analysis Branch

National Oceanic and Atmospheric

Administration

U.S. Department of Commerce

Federal Building

Asheville,NC 28801

704-271-4765; Fax 704-271-4246

E-mail: mchangery@ncdc. noaa.go

v

Dr. C. Y. Chen

Senior Civil/Geotechnical Engineer

Office of Foreign Buildings

Department of State

Code SA-6, Rm. 347

Washington, DC 20520

703-875-6207; Fax 703-875-6204

E-mail: chency(^state.gov

Mr. Vincent P. Chiarito

Research Structural Engineer

Structural Mechanics Division

Structures Laboratory

U.S. Army Engineer Research and

Development Center

3909 Halls Ferry Road

Vicksburg,MS 39180-6199

601-634-2714; Fax 601-634-3412

E-mail: chiariv@exl .wes.army.mil

Mr. James Lander

Geophysicist

Cooperative Institute for Research in

Environmental Sciences

University of Colorado

Campus Box 449, Room 152 RL3

3100 Marine Street

Boulder, CO 80309

303-497-6446; Fax 303-497-65 1

3

E-mail: jlander(^ngdc.noaa.gov

Mr. Michael Mahoney

Physical Scientist

Federal Emergency Management Agency

500 C Street SW

Washington, DC 20472

202-646-2794; Fax 202-646-4387

E-mail: mike.mahoney@fema.gov

Dr. Martin C. Miller

Chief, Coastal Oceanography Branch

U.S. Army Engineer Research and

Development Center

Coastal and Hydraulics Laboratory

3909 Halls Ferry Road

Vicksburg, MS 39180

601-634-3999; Fax 601-634-4314

E-mail: m. mi1 ler@cerc . wes. army.mil

Dr. Erdal Safak

Research Structural Engineer

U.S. Geological Survey

Box 25046, MS 966

Denver Federal Center

Denver, CO 80225

303-273-8593; Fax 303-273-8600

E-mail: safak@usgs.gov

30
Dr. John B. Scalzi

Program Director

Civil and Mechanical Systems Division

National Science Foundation

4201 Wilson Boulevard, Room 545

Arlington, VA 22230

703-292-8360; Fax 703-292-9053

E-mail: jscalzi@nsf.gov

Resolutions

33

RESOLUTIONS OF THE THIRTY-SECOND JOINT MEETING

U.S.-JAPAN PANEL ON WIND AND SEISMIC EFFECTS (UJNR)

National Institute of Standards and Technology, Gaithersburg, MD

16-19 May 2000

The following resolutions are hereby adopted:

1. The Thirty-Second Joint Panel Meeting provided the forum to exchange valuable

technical information that is beneficial to both countries. In view of the importance of

cooperative programs on the subject of wind and seismic effects, the continuation of Joint

Panel Meetings is considered essential. Both sides agreed to explore new ideas and areas

to strengthen cooperative activities in support of the Panel's mission.

2. The following activities have been conducted since the Thirty-first Joint Meeting:

a. Technology Exchanges. Technical experts, technical documents, and applications

of electronic media have been exchanged. These exchanges have contributed to the

development of new research programs and enhanced ongoing research in both countries.

b. Task Committee Workshops. The Panel held five workshops, symposia, or committee meetings; approximately 300 specialists from both countries participated in

these activities:

1. Task Committee (C), 6th U.S. -Japan Joint Technical Coordinating Committee

Meeting on Composite and Hybrid Structural Systems, 25 March 2000, Los

Angeles, USA.

2. Task Committee (C), 1st U.S.-Japan Technical Coordinating Committee Meeting on Auto-Adaptive Media (Smart Structural Systems), 6-8 January 2000, Tsukuba,

Japan.

3. Task Committee (E), 1st U.S-Japan Workshop on Design for Wind and Wind

Hazard Mitigation, 24-26 May 1999, Tsukuba, Japan.

4. Task Committee (G), 1st U.S. -Japan Workshop on Seismic Information Systems,

15-16 November 1999, Tsukuba, Japan.

5. Task Committee (J), 15th Bridge Workshop, 9-10 November 1999, Tsukuba,

Japan.

c. Common Agenda. The Panel recognizes the importance of providing earthquake

science and technology expertise to the 17 April 1996 U.S.-Japan Natural Disaster

Reduction Initiative of the U.S. -Japan Framework for New Economic Partnership

35
(Common Agenda). Panel members organized and participated in the First HighLevel U.S. -Japan Earthquake Policy Cooperation Forum, 20-22 October, 1998,

Seattle, Washington, USA, and in the Second Forum, 10-12 November, 1999,

Yokohama, Japan. The Panel will continue to provide, in close cooperation with the

Earthquake Disaster Mitigation Partnership, technical leadership and participation in

the Third High-Level Forum Meeting that will address five themes: 1. Strategies to encourage implementation of preparedness and loss prevention measures; 2. Exchange of information and experience on tsunami hazards and risk reduction; 3. Lessons learned from recent large-scale earthquakes around the world; 4. Establishing the development of a joint presentation that describes the High-Level

Forum achievements and a strategy for international dissemination; 5. Plans for

future cooperation. Appropriate Task Committees will respond to the Common

Agenda in planning their work.

3. The Panel approved the Task Committee reports presented during the 32nd Joint Panel

Meeting. Each report included objectives, scope of work, accomplishments, future plans,

and other information. The Panel will continue to review the Task Committees' progress

toward meeting their objectives; consolidating, eliminating, and/or creating Task

Committees as desirable.

4. The Panel endorses the following five (5) proposed Task Committee Workshops or Committee Meetings during the coming year:

a. Task Committee (A), 2nd U.S.-Japan Workshop on Soil-Structure Interaction, March,

2001, Tsukuba, Japan.

b. Task Committee (C), 2nd U.S. -Japan Technical Coordinating Committee Meeting on

Auto-Adaptive Media (Smart Structural Systems), by the end of 2000, USA.

c. Task Committee (C), International Workshop on Performance-Based Design of

Building Structures, November 13-15, 2000, Tsukuba, Japan.

d. Task Committee (D), 3rd U.S.-Japan Workshop on Advanced Research on

Earthquake Engineering for Dams, in 2001, USA.

e. Task Committee (G), in coordination with Task Committee (A), Workshop on

Seismic Information Systems, 2000, in conjunction with the 3rd High-Level Forum

- " Meeting, USA.

Task Committee (I), 6th Tsunami Workshop, 200 1 , USA.

Task Committee (J), 16th U.S. -Japan Bridge Workshop, October 2000, Lake Tahoe,

USA.

In the event that T/C co-chairs deem it essential to conduct a joint meeting or workshop

36
that is not included in this endorsement, the T/C co-chairs will make such a request through

their respective Secretary-General for approval by the U.S. and Japan Joint Panel

Chairmen.

Scheduling for the Workshops will be done by the U.S. and Japan Chairmen of the Task

Committees with concurrence of the Joint Panel Chairmen. Both sides' Secretaries-General

will be kept informed of the planning, conduct and results of workshops or committee

meetings including resolutions and reports. Results of each activity conducted before the

33rd Joint Panel Meeting will be presented at the 33rd Joint Panel Meeting.

5. The Panel recognizes the importance of continued exchange of personnel, technical

information, research results, and recorded data that lead to mitigating losses from strong

winds and earthquakes. The Panel also recognizes the importance of using available largescale testing facilities and complementary capabilities in both countries. Thus, these

activities should be continued, strengthened, and expanded. Further, the Panel will study

the feasibility of establishing a network for live transmission of experimental testing,

including data, models, and simulation tools.

6. The Panel recognizes the significant vulnerability of residential construction to natural

hazards and believes that reduction of life and property losses in the housing sector is

critical for both sides and in other parts of the world. The panel affirms the need to make

disaster-resistant housing a priority and to strengthen activities in support of this objective

in the future.

7. The Panel recognizes the importance of continuing its joint research programs on Soil

Liquefaction and Countermeasures and on Auto-Adaptive Media (Smart Structural

Systems).

8. The Panel recognizes the importance to continue to seek means to contribute to the

International Organization for Standardization (ISO) through participation of its members

and related agencies in appropriate ISO Technical Committees. The Panel also recognizes

the importance of continuing the related information exchange to contribute to the ISO.

9. The Panel will continue to transfer technical information obtained through the Panel and

Task Committee activities including technology transfer on wind and seismic effects to developing countries.

10. The Panel encourages all members from both sides to use electronic media, as much as

37
possible, in communicating Panel findings and summaries of its activities. The Public

Works Research Institute (PWRI) and the National Institute of Standards and Technology

(NIST) will maintain the electric mailing list aid in the distribution of information to Panel

members.

1 1 . The Panel, in recognition of its many strengths and past successes and a desire to position

itself both functionally and organizationally to address future challenges, will develop

jointly a strategic Plan for the UJNR Panel on Wind and Seismic Effects. An Ad-Hoc

Committee will be formed to develop the Strategic Plan that will:

Evaluate the current organization, task committee structure, annual technical meeting

and site visit program, cooperative research projects, and related operating procedures.

Assess current and emerging technical needs and interests of government agencies in

the two countries such as Performance-Based Design and Rehabilitation.

Review options for Panel affiliations/linkages with different U.S.-Japan governmentto-government cooperation mechanisms, including participation of industry and

university representatives.

Review options for streamlining and increasing the efficiency of Panel operations and

for moving toward more effective operating procedures.

Develop new proposals for Panel functional statement, organizational structure and

membership, operational procedures (including procedures for requesting and

approving the formation of new Task Committees and for conducting cooperative

research projects and workshops), and government-to-government affiliation

mechanisms.

Additionally, the Ad-Hoc Committee will take into consideration recommendations from

the report on U.S. -Japan Dialogue on the Role of Science and Technology in Society into

the New Millennium in forming its recommendations. The U.S.- and Japan-side Chairs

will nominate two individuals each to serve on the Ad-Hoc Committee as soon as

possible. The Ad-Hoc Committee will present the Draft Strategic Plan by October 1, 2000

for joint review and approval by the U.S. and Japan-side Chairs.

12. The Panel published its seventh newsletter issue, Wind and Seismic Effects, Winter 2000.

Continuation of the publication of the Panel's newsletter in accordance with the Panel

Strategic Plan.

13. The Thirty-Third Joint Panel Meeting of the UJNR Panel on Wind and Seismic Effects

will be organized by the successor to the Public Works Research Institute, Tsukuba,

Japan, after the reorganization of the Japanese Governmental Ministries and Agencies, in

May 2001. Dates, program, and itinerary will be proposed by the Japan-side with

38
concurrence of the U.S. -side Panel.

39

Papers

41

Theme 1

Information Technology for Land Management

and Natural Disaster Prevention

Countermeasures

43

Web-Based Geographic Information System Support to Disaster Response

William E. Roper'

Abstract

Natural disasters have a major impact, globally and within the United

States (U.S.) causing injury and loss of life, as well as economic losses.

To better address disaster response needs, a task force has been established

to leverage technological capabilities to improve disaster response

management Web based geospatial analysis is one of these important

capabilities. Samples of geospatial technologies applicable to disaster

management are presented. These include 3-D visualization,

hyperspectral imagery, LIDAR, use of spectral libraries, digital

multispectral video, radar imaging systems, photogeologic analysis and

geographic information systems.

Key words: Data management; Disaster management; Global Disaster

Information Network, mapping, hyperspectral; web mapping, geospatial

library, LDDAR, radar imaging, GIS, Global Positioning System, satellite sensors, 3-D fly through.

1. Introduction

Despite technological advance, disaster

risk continues to grow. Emergency

managers and others continue to be

called on to make decisions during

disaster events, as well as in the pre-andpost disaster phases, with incomplete

information. In order to make optimal

decisions to reduce the loss of life and

property, stakeholders uniformly must

be able to obtain the needed information

in a format that is appropriate for their

capabilities. (NRC 1998).

This paper is intended to stimulate the

systematic consideration of the

information technology and information

processing requirements associated with

the larger community of Disaster

Management. This document has two

audiences. For technologists this

document is intended to provide

background and overview material

pertinent to the disaster community at large. For disaster stakeholders this may

serve as an introduction to distributed

interoperable technologies and the role

of scenarios in developing appropriate

solutions.

Natural disasters are a constant threat to mankind on a global scale. Global

disaster costs are continuing to rise. Annual global economic costs related to

disaster events average $440 billion per

year (World Disaster Report, 1 996) with

floods being the major cause. In the

U.S., the number of lives lost due to

natural disasters has been decreasing over the last several deceased, largely

because of advances in disaster

indication and warning capabilities. In terms of damage to property, however,

the trend is reversed. For the period of

1992-1996, the average cost of natural

disasters in the U.S. has been $54.3

billion, with hurricanes and earthquakes

45
tied as the leading cause. These rising

costs are the combined result of

increased urbanization, particularly in high-risk coastal areas, and the increased

complexity and size of our

infrastructure.

The loss of life and property continues to

rise in many regions of the world

because of these events. One example is the Bangladesh weather event of 1970,

when a tropical cyclone slammed into its delta region killing 300,000 people

(Tobin and Montz, 1997). The crop

losses were estimated at $63 million, and

more than 280,000 cattle were drown

(Burton, Kates, and White, 1993). The

rich delta soil is an agricultural resource

that still draws people to settle there;

therefore a recurrence of this type of

weather event could likely pose a similar

disaster. Another more recently example

was Hurricane Mitch in the fall of 1998

that resulted in over three thousand

deaths and massive damage in Central

2. Data Management for Disasters

It is clear that despite excellent efforts

by many groups the approach to providing information for disaster

management is not effectively utilizing a wealth of data that resides, with various

organizations and that existing

technology could deliver to disaster

mangers important information products

that could save lives, reduce damage to property, and lessen the environmental

impacts of disasters. The current

situation is characterized by numerous

shortcomings that inhibit optimal

decision making for disaster

management. The inability to access information and the lack of

standardization, coordination, and

communication are all obstacles that

need to be overcome [NRC 1998].

Considerable effort has been expended

throughout the Disaster Management

community to articulate issues and to characterize the dynamics and the interrelationships that need to be

accommodated in a viable information

management strategy An important

guide for this effort is found in the

recommendations made by the Board On

Natural Disasters (BOND) in their report

to the National Research Council (NRC

1998). The BOND's primary goals as

articulated in its cha were to: • Improve decision making before,

during, and after emergencies

through better access to and

quality of data and information

• Identify users and their needs

• Provide information products

specifically designed to meet

users' need

• Promote efficiency and cost

effectiveness

• Stimulate and facilitate

mitigation

2.1 Board on Natural Disaster

Observations and Recommendations

The BOND responded to its charter with

a series of observations and

recommendations that provide

significant input and guidance to the

formative process needed for improved

disaster management. The BOND in its report put the following observations

forth.

• The need for an improved

information network and its potential benefits are clear

• The foundations of an

information network are already

in place.

46
• Recent advances in technology

provide the mechanism for

establishing a network.

• Successful implementation of the

disaster information network

concept will require a commitment of resources from a broad spectrum of stakeholders. In addition to these observations the

BOND provided a set of

recommendations to the community at large. These recommendations are consistent with much of the philosophy

that drives this initiative They also help

to forge a realistic set of parameters and

directives that can be used to focus on

the disaster management community.

• The products of Disaster

Information Network should be

based from the outset on user's needs.

• A major of a Disaster

Information Network should be on integration of various data

types.

• Data and information quality and

reliability are major issues that

need to be addresses.

• An effective dissemination and

access plan is critical to the success of any information

network.

• The Global Disaster Information

Network Transition Team should

focus initially on establishing a

national disaster information

network (i.e., with a U.S. focus),

but the model should be extended

to a global process (Global

Disaster Information Network) as soon as it can be demonstrated

that a Global Disaster

Information Network is technically and organizationally

feasible (GDIN 1997).

In addition, to the BOND discussion of

focus areas for improvement and growth

there are a number of functional areas

related to data and information

processing The four main areas for

improvement in information

management are: • Locating information

• Assessment and evaluation of

information integrity, currency

and accuracy

• Assimilation and Integration of

new information

• Application(s) that incorporate

new information to produce

customized, understandable

(interpretable) domain specific

(context specific) results

BOND conclusions were weighted

towards developing integrative products

to assist in decision support and

planning. For instance, a series of

recommendations that emerged from the

NRC publication included:

• International, national, regional, or local maps showing how

hazards and risks vary in space

and time;

• Estimates of probability of

occurrence of hazardous events,

• Estimates and examples of

potential effects, especially for

structures;

• Real-time display of what is happening during the course of a

disaster,

• Systems for contingency

planning; and

• Codes, standards, and

construction methods for

structures. These are important components in any

workable framework for comprehensive

disaster management. However, the

focus here is directed towards derived

products, procedure and protocol. It is

47
important that the process develop

baseline products and implement

solutions that reflect the complex fabric

of the disaster management community.

This work is done at a lower, more

formative level in the information

management life cycle and in fact forms

the basis for the production of the

products referred to by the BOND. As

noted above the assessment process will

work to identify the constituent

components associated with specific

disaster events and scenarios. The

constituent components will then

become the basis for technology

specifications and the delivery of

capabilities that encompass the products

identified by the BOND.

2.2 Disaster Data

Considerations

Management

This section provides assumptions based on information and conclusions that

have emerged from recent studies in disaster management community itself (NAPA 1998). These assumptions

certainly do not constitute and

exhaustive set of parameters. Rather,

they provide a point of departure and a

set of "operating hypotheses" that can be

used to help frame discussion and

consideration of needs and requirements.

2.2.1 Information Availability

Available Information and capacity is not uniform, consistent and standardized

for disaster managers. Large cities may

have invested millions of dollars in their

information systems, while smaller

municipalities may be operating with one personal computer.

2.2.2 Multiple Boundary Problem

Disasters will constantly crosscut

established boundaries. A fundamental

problem in dealing with disasters is that

they do not respect boundaries that

include organizational, political,

geographic, professional, topical, or

sociological consideration.

2.2.3 Coordination & Communication

Challenge

Disasters will often overwhelm

mechanisms for coordination and

communication. In addition, the

mechanisms to bring data and

information to decision makers are uncoordinated Information is often

produced from disparate sources and

transmitted in whatever format the

provider prefers, requiring significant

effort to compile it into a form that

provides a coherent picture or even

thwarting integration altogether. Data

standards are often inconsistent, and, even more dangerous; users are sometimes unaware of the limitations

and uncertainties in data or are presented

with conflicting interpretations of data

without the means to assess the

reliability of the sources. All of these

issues reduce the efficacy of the

decision-making process (NAPA 1998).

The problem is compounded because

information delivery systems in many

cases become overloaded during crises.

For instance, in 1996 a moderate

(magnitude 4.7) earthquake in San Jose,

California, led to more than a million

attempted hits in less than one day on the

U.S. Geological Survey's World Wide

Web earthquake information server. Most of those attempts were

unsuccessful, including those by

emergency managers trying to access

data and information to aid in the

formulation of a response plan.

48
3. Disaster Management Process

Issues

Disaster management is an exercise in logistics and information processing and

distribution. To effectively undertake

these tasks requires a good

understanding of disaster information

requirements and the characteristics

associated with them Disaster events are not all created equally. They come

indifferent sizes, have different behavior

and can be categorized on the basis of

their impact on natural resources,

agriculture, communities etc. They can

also be discriminated and classified

along a number of dimensions including

impact, severity, duration, geographic

setting and advance warning. In order to develop an Information Technology

architecture or even plausible use-case

scenarios it is essential to understand

the disaster event from the perspective of

those responsible for assimilating the

data and information and producing

operational plans to deal with the

disaster event and its aftermath.

For instance, it is important to appreciate

the four basic phases of disaster

management may actually manifest

themselves in different ways depending on the nature of the disaster. The

mitigation phase is essentially a continuous process of preparation,

analysis of performance and subsequent

modification and refinement. As a result

it will likely exhibit the smallest amount

of variance. However, the other phases

(preparedness, response and recovery)

will likely have significantly different

attributes depending on the nature of the

event. For example consider the

difference in the preparedness phase that

would be associated with major seasonal

flooding along the Mississippi or the

Missouri Rivers, the Chaparral fires along the Pacific Coast, a Tsunami that

impacts much of the Pacific Rim and an earthquake such as Northridge or Loma

Prieta (Roper 1999). In each case the

geographic coverage and the amount of

time for planning and staging is very

different. As a result, many of the other

systemic aspects of the disaster

management initiative will be different as well.

An event warning which impacts the manner in which the preparedness phase

is managed also has implications and

consequences for the recovery phase

(e.g. In the case of the Loma Prieta

earthquake the emergency response

resources were directed to the bay area

despite the fact that Santa Cruz and

Watsonville had more significant

damage). Partially as a result of this

California is building the Tri-Net

seismic sensing network to measure

shake and model damage in order to make more informed decisions regarding

staging recovery resources.

4. Disaster Management Communities

The disaster management is extremely

heterogeneous. In order to develop

effective systemic solutions to the needs

of this community any framework must

address the need to effectively and

efficiently share information across

institutional boundaries within and

between the groups that make up this

community. In essence an interface

must be specified at each boundary that

separates levels in an agency or at the

juncture between different agencies or between different sectors. These

interfaces each have at least three

components: 1) technical, 2) institutional

49
and 3) institutional administrative. The

importance of these interfaces is also

proportional to the difficulty of

implementing them in an automated

context. For instance, the lightest weight

interfaces should be between levels

within the same entity within a single

sector. In contrast the heaviest

interfaces would be those that attempt to span sectors. A number of these communities have

made significant institutional

investments in the construction of

foundation data products that feed

modeling, analysis and decision support

in a range of contexts. Not surprisingly

the situation in the disaster management

community in regard to data parallels

spatial data issues across the federal

government and throughout state and

local government and the private sector.

The issue of data sharing by the disaster

management community was addressed

by the BOND. Their findings

recognized that despite the importance

of significant databases, their utility is impaired by a host of problems deriving

from incompatible formats, inconsistent

geographic reference systems,

conflicting standards, and other humancaused factors. Many of these problems

could be resolved and the value and

utility of the databases for disaster

decision making greatly enhanced

through improved organizational and

technological coordination with only an

incremental increase in cost. This is clearly in the public interest to accomplish this (NRC 1998).

Disaster management scales for

international agencies and governments

to the individual. However, a very

significant level of responsibility is

vested at the local level. For instance, it

is the on the ground officials, many of

whom are local, that ultimately play the

role of first responders (e g police, fire,

medical and public works) and who also

manage many of the recovery

operations.

To be effective in this context protocols

must define communications, database

structure, data formatting, hardware and

software requirements, networking,

quality control, and other issues needed

to assure the linkage will improve

decision making before, during, and

after emergencies through improved

access to quality information. In

addition, such a linkage will provide

information products that are specifically

designed to meet the needs of users

5. Common Characteristics of

Disasters

Disaster management is not a linear

process that can be documented easily in a flow chart with a readily apparent

beginning and absolute end point.

Rather, it is a cyclical process of

approximation, response and re- calibration that involves many different

actors whose roles in relation to one

another are likely to change based on

circumstances and the stage in the

process. The one constant evident in the

process is the chaos that drives the

system. Another absolute is that the

effective, efficient application of

information technologies and products

has the capacity to improve the system

in a number of ways that will combine to save lives, mitigate overall damage,

conserve resources and ameliorate

human suffering.

50
To develop effective architectures and

technologies that meet the needs of the

disaster management community there

must be a precise understanding of the

disaster management Hfe-cycle

(Mitigation, Preparedness, Response and

Recovery), the Information

Communities that combine to define the

disaster management community and the

information processing requirements

associated with the cycle of data

development, dissemination, analysis

and review. In addition, and perhaps

more importantly, there must be a precise understanding of the dynamics

between these components and the

"interfaces" that these dynamics imply.

Only with such an understanding can we

effectively model the process and derive

technology solutions that map well into

the business model of disaster

management.

6. Data and Information Processing

Requirements

Data needs, the characteristics of data

and the ability of data to contribute to

analysis, response and recovery is well

understood. The basic attributes of

information that are considered most

important by this community include:

• Timeliness: delivery of data and

information in time to drive

decision making

• Consistency: delivery of data and

infonnation in a consistent,

uniform manner

• Understandability: delivery of

data and information in a manner

that is appropriate and

understandable in the target

community

• Accuracy: precision in measurement and observation

• Flexibility: adaptability to multiple situations

These information attributes have

evolved over time within the disaster

management community representing

the viewpoint of a range of providers, users and consumers scaled from the

local to national level In this regard the

communities involved transcend

boundaries between recognized sectors

at each of these levels (private, public,

not-for-profit, NGO etc.).

7. Hurricane Scenario

Hurricanes, cyclones and tropical storms

are among the most visible disasters that

visit coastal areas around the world on a

regular, seasonal basis. They are

familiar to most everyone and like other

kinds of disasters strike fear into those

who have come into direct contact with

them.. In the remainder of this paper a hypothetical scenario will be developed

for the response to a hurricane drawing

attention to the multiple actors and

information communities that are involved and the potential for highly

heterogeneous information and operating

environments that may exist during such

an event.

7.1 Hurricane Disaster Description

Hurricanes generally emerge as tropical

depressions in the June to November

time frame. The primary forces at work

in a hurricane are high, sustained winds;

coastal flooding firom storm surge;

inland flooding from heavy rains and

battering and beach erosion from

heightened wave action. There are countless secondary hazards as well

including tornadoes, other severe storms

and infrastructure failure (e.g. seqage

treatment, power plants etc.). High

51
winds pose significant structural load

risks as well

Hurricanes are "ranked" on the SaffirSimpson scale from 1 (74-95 mph

sustained winds) to 5 (greater than 156

mph sustained winds) From a ranking

three and higher (winds greater than 1 1

1

mph) large trees can be blown down.

Structural damage will also occur,

particularly for less substantial

buildings. Strom surge of between 9 to

1 8 feet can be expected which will result

in coastal flooding and damage to natural and man made structures. Waves

of 18 to 25 feet will be possible and will cause significant erosion and damage to shoreline property.

7.2 Event Timeline

A hurricane event timeline is dehmited

in phases that correspond to the onset of

the storm measured in hours. This

system covers pre-storm, landfall and

continues until the completion of

response activities. The Federal

Emergency Management Agency

recognizes three primary phases.

• Awareness; 72 to 60 hours before

the arrival of gale force winds

(32 to 63 mph)

• Standby: 60 to 48 hours before

the arrival of gale force winds. It

is likely that a tropical storm

watch would be issued during

this period.

• Response: 48 hours before arrival

of gale force winds through

termination of the emergency.

Hurricane watches and warnings

would be issued by the National

Weather Service during this

period

Preparing for an approaching hurricane,

response and recovery requires

incorporating meteorological data into

models that predict storm track and

storm surge These predictions are then

incorporated into maps showing

population distribution, evacuation

routes, deployment of emergency

personnel and supplies, and other

relevant response information. This

effectively brings to bear the collective

capabilities of many federal, state and

Local agencies and requires the support

of a number of private sector entities that

maintain basic infrastructure (utilities,

tax base, insurance) and communications

media. In addition to federal agencies

there are a number of officials at the

local level that act in a "hands-on" manner throughout the emergency. In essence, many of the participants and

their domains at the local level are characterized in Table 1 Many of these same types of participants would be

reflected at the state level also.

The matrix in Table 1 has been

annotated with the Federal Emergency

Management Agency's assessment of

lead agencies and departments. In some

cases an agency or department will play

both a lead (P for primary) and a secondary (S for secondary) role

depending on the scope etc. of the event. For the purposes of this example each

cell boundary in the Table 1 matrix can be thought of as a place holder for one or more interfaces that will be required in a

final, fielded application for disaster

management. However, once the

participants are identified the

relationships and dependencies between

these groups must be described in

precise event by event detail. When this

has been accomplished the participants

from many of the communities will

come into play and their roles,

52
responsibilities and mandates must be

integrated seamlessly with this matrix.

7,3 Information Needs

The information requirements associated

with the disaster life cycle, hurricane

scenario, are not insignificant. By rough

count, a minimum of forty separate

entities spread across the continuum of

providers and consumers that extend

from the federal to the local level and

cross well established boundaries that

exist between the public and the private

sectors must work together in a seamless manner throughout to ensure success. These communities have diverse

expectations, requirements, technical

capabilities, missions, different hardware

and software etc.

The process must define the

requirements and expectations of each

community and develop an appreciation

of the input they expect and the output

they produce in order to begin to conceptualize the nature and the extent

of the interfaces that are required to meet

the needs of this community. However,

a major issue is the need for

development of good schemas that have

been successfully federated across these

information communities. It is also clear

that a range of data products (raster,

vector, video, voice etc.) will need to

fielded across a range of platforms and

harmonized with users who require tools

for analysis, decision support and

mission planning.

8. Web Based Geospatial Test-Bed

Demonstration of Hurricane Scenario

The Web Mapping Technology Program

is a collaborative effort between military

and civil agencies of the US

government, international organizations,

and the commercial geo-processing

technology sector to develop

interoperable web mapping technology

solutions. The program will advance the

current state of proprietary web mapping

technology solutions, enabling webbased mapping to support diverse

applications that require access to

multiple, distributed geospatial

information sources across the World

Wide Web (Figure 1). Applications for

this kind of interoperable technology

include environmental analysis and

management, disaster relief operations,

emergency response and management,

and military operations.

The Test-bed will identify and develop

software components implementing

common interfaces and protocols

necessary for web-based geospatial

information display and interaction

services. These components will support

diverse appUcations across the spectrum

of web user environments, ranging from

enterprise intranets to home users on

dial-up connections. Based on the user environment and potential applications,

the Test-bed design acknowledges that

there will be at least three different cases of client-server distribution of webbased display and interaction services.

The cases are distinguished how much

actual geospatial information,

symbolized map displays (i.e. gif or jpeg

images), or intelligent graphics travel across the web and the client-server

distribution of services necessary to support that transfer. The Test-bed

design process also anticipates that the

cases can also be combined. A major

focus of the test-bed will be to assess the

content, format, and transport

mechanisms of the web communications

necessary in each case. The test-bed will

also explore methods to interlink

standards-based, geospatial information.

53
allowing nonlinear discovery and

integration of distributed geospatial resources by information use and

content. The effectiveness of these

architectures will be tested by

conducting scenarios to assess the

capability of interoperable web mapping

to support environmental management,

emergency management, disaster relief,

and military humanitarian relief and

stabilization operations.

Early in the Web Mapping Test-bed

(WMT) project, two fictional scenarios were created to generate requirements

for data discovery, access, and display,

and to keep the demonstration "focused"

and centered on "real world" issues. One

scenario centered on the needs of a TV

journalist who uses the web to access

spatial data to help her create a news

report about a hurricane approaching

Mobile, Alabama. In the demonstration,

the journalist uses a "paper thin"

desktop, it merely needs to run Netscape. The other scenario focused on

the needs of a team of disaster managers

in a mobile disaster response control

center, preparing for the same hurricane.

This team has additional software,

including "pre-built" queries to generate

appropriate views. In both cases, online

geo-data servers were registered in a Open-GIS Catalog Specification catalog

that enabled a query very much like an

ordinary web search engine query.

In the first scenario, the journalist using

a standard web browser was presented

with a "shopping basket" metaphor that

allowed browsing geospatial data sources of potential interest (figure 1)

The journalist selected a data source and

specified a region to view. The zoomed

map or image came into view on the web

browser, just as one can see now with

today's non-interoperable web mapping

applications The difference from

today's technology became clear when

the journalist selected 4 additional sources and added them to the "shopping

basket." Now the standard web browser

was used to access these 4 additional

spatial data servers, and these servers provided data that automatically overlaid

the first map (figure 2). Five layers were

displayed, each a live "view" into data on separate servers around the world

(figure 3). The servers were from

different vendors. Some served vector

data of different kinds, others raster data.

Any combination of map layers could be

viewed, overlaying each other in any

stacking order (figure 4). Zooming on

one map or image caused the various

servers around the world to perform

identical zooms, so that the layers would

stay identically geo-referenced (figure

5). The maps and images were not stored on the servers as gif images, but they were created as requested, and delivered

to the web browser as gif images.

In the second scenario, more capable

clients, GISs and custom software from

different vendors, were used to access some of the same servers used in the first scenario. These clients created smarter

displays using previously built and

stored presentation styles (figure 6). For

example, as shown in figure 7 storm

surge information from one source, and

evacuation route information from

another source could automatically be

combined to highlight routes cut off by

the storm using a single "pre-built

button."

The relatively simple new web mapping

protocols and interfaces which enable

the interoperability described above

depend on foundation work done earlier

54
in Open GIS Consortium This includes

the work underlying the Open-GIS

Simple Features Specification and OpenGIS Grid Coverage Specification. New

Open-GIS Specifications based on these

new protocols and interfaces are now

approved as Open-GIS Specifications.

The Open GIS Consortium's software

vendor members are expected to implement these interfaces in client and

server products which will interoperate.

As with all Open GIS Consortium

specifications, the Web Mapping

protocols will be made available free on

their web site to all software developers

within a few months

There were essentially two types of

applications: Web Mapping CUents, and

Web Mapping Servers. The Clients

created requests that conform to the Web

Mapping Protocols. The main interface

is the Get-Map protocol, that identifies one or more layers to be displayed, along

v^th the various parameters needed to ensure the display is exactly the one

intended. Examples of these parameters

are the Spatial Reference System of the

display, the center and dimensions of the

display window, instructions for

transparency, the order in which to display layers, and so on. Two

additional interfaces are provided: GetCapability and Get-Feature-Info. These

protocols enable a client to discover the

abilities of a server, and the ability to

"drill down" through a display to discover the attributes of a displayed

feature.

The Web Mapping Servers are able to

interpret requests that conform to the

Web Mapping Protocols and able to generate appropriate objects to be

returned to requesting clients.

Additionally, a "Cascading Server" was

used that is able to query other servers and report their capabilities as if those

capabilities were its own. The

Cascading Server invokes other servers whenever a cascading service is requested.

8.1 Demonstration Data Providers

A number of participating agencies and

private sector organization provided data

and software for this demonstration.

The following providers were part of the

demonstration: City of Mobile GIS

Department, Raytheon ITSS (STX),

FEMA, NOAA/NESDIS, NRLMonterey, NASA-JPL, USDA-NRCS,

USATEC, Alabama Coastal Hazards

Assessment CD-ROM, NOAA Coastal

Services Center in partnership with

Region IV FEMA, Alabama Emergency

Management Agency, Alabama Dept. of

Economic and Community Affairs, and

South Alabama Regional Planning

Commission

8.2 University and Private Sector

Participants

Participating university and private

sector technology providers included:

Autodesk (US), BBN (US),

Technologies Blue Angel Technologies

Cadcorp (UK), Compusult (Canada),

Cubewerx (Canada), ESRI (US), Galdos

Systems (Canada), Geodan IT

(Netherlands), Geomatics (Canada),

Hitachi (Japan), ILOG (France),

Intergraph (US), Ionic Software SA

(Belgium), Laser Scan (UK),

Litton/TASC (US), Lockheed-Martin

(US), Maplnfo (US), Microsoft (US),

MIT (US), Mitsubishi (Japan), NTT

Data (Japan), ObjectFX (US), Oracle

55
Corp (US), Ordnance Survey (UK), PCI

. Geomatics (Canada), Penn State

University (US), SICAD Geomatics

(Germany), Social Change Online

(Australia), Sun Microsystems (US), and

Universal Systems (Canada).

9. Conclusions

The Web Mappmg Test-bed is an

opportunity for technology,

infrastructure, and geospatial

information providers and users to mutually define interface and protocols

: (and thus candidate interoperability

specifications) in the context of a hands- on engineering experience expected to

10. References

Burton, Ian, Robert W. Kates, and Gilbert

F, White, The Envirnment as Hazard, 2d

Edition, New York, NY: The Guilford

Press, 1993

Disaster Information Task Force Report,

Harnessing Information and Technology

for Disaster Management, Global Disaster

Information Network, 1997

NAPA Report, Geographic Information for

the 21"* Century: Building a Strategy for the

Nation, National Academy of Public

Administration Press, 1998

shape the fijture of Web mapping

software development and Web data

publication The Web Mapping

Technology Initiative is expected to generate a significant capability and

support a long-term effort. Current and

fijture work efforts include developing

Map-Server interface specifications,

XML encoding standards for geospatial

information, client interactivity and

discovery tools, standards for geospatial

hyper-linking, and standards-based

methods to encode and transport

information about spatial-temporal

events such as floods, earthquakes and

hurricanes

NRC Report, Information Infrastructure for

Managing Natural Disasters, Board on

Natural Disasters, National Academies

Press, 1998

Roper, William, Geospatial Technology

Support to the Nation's Navigation System,

Transportation Research Board, National

Research Council, Washington, DC,

January 1999

Tobin, Graham A., and Burrell E. Montz,

Natural Hazards: Explanation and

Integration, New York, NY: The Guilford

Press, 1997

56
Organization

Type

Direction

And

Control

CommunicationsWarning Emergency

Public

Information

Evacuation

Mass

Care

Health

And

Medical

Resource

Management

CEO P

Fire P/S

Police P/S

Health &

Medical

P/S P

Public

Works

P/S

Emergency

Programs

EOC

Manager

Communictions Coord.P

Public

Information

Officer

P

Evacuation

Coordinator

P

Mass Care

Coord.

P

Resource

Manager

P

Education

Dept.

Animal

Control

Warning

Coord.

P

Volunteer

Organizations

Other

Organizations

Table 1 : Public Sector Organizational Responsibilities in Response

57
Figure 1 : Examples of world wide maps, images and information sources

Figure 2: Geospatial spatial sources including new high resolution Ikonos commercial

imagery

58
j-Source Gl Visualization

• EnglTn

Conceptual Architecture

EynamiQ

User Profile Registry Applications

Itublic Safety

Users

Provide rapid updates to the

enterprise node

\ / \ ^ ^ \ / Census \ / \

/ Data \ / WeatherX

ay

and Integration Service:

j WMT Discovery Services

Integration Tools

Security Functions

Optiona

Sar iple View Services for Integration

Firew/all

pervide

itry

Figure 3 : Web based server structure and interaction capabilities

Figure 4: Interactively developed products being used by emergency decision makers

59
New Model for Seodata Use

Integrated View

Queries

extract data from

diverse sources Internet Common tterfaces enlble interoperability p- " "j

Data^^ndor City Agency State Agency Fed. Agency

: Metfldata,^ Data ^ Data ^ 1. ^MetodotjaMl

Data

<; A\etad(^H

CleadMhouse

Catalog that

indexes data,

similar to WWWs

html search engines

Figure 5; Multiple data inputs of different type used to create integrated products

Figure 6: Integrated overlay and scale change for products from world wide data

60
Figure 7: Application of high resolution commercial satellite imagery into products

DRAFT MANUAL FOR SEISMIC INFORMATION SYSTEM DEVELOPMENT

by

Hidekj SUGITA'\ Tomofumi NOZAKl^^

ABSTRACT

This paper introduces the "Draft Manual for Seismic Information System Development" that Public Works Research Institute, Ministry of

Construction recently developed. The manual was prepared to develop efficient seismic

information system: SIS for civil infrastructure

administrating bodies from the viewpoint of the

national land management policy. This paper

includes a basic concept of SIS development,

requirements of SIS, element technologies for

SIS, and safety/ security.

KEY WORDS: Earthquake Disaster

Information System

System Requirement

Safety, Security

1. INTRODUCTION

Recent situations such as environmental changes

and severe disasters over the world require more

comprehensive countermeasure than before. Such

countermeasure must be an overall program

including monitoring of circumstances, adequate

planning and administration in addition to the construction. Especially, civil infrastructures

require overall management because they have

counteractions to nature and social activities.

The Ministry of Construction has initiated a

five-year comprehensive research project

"Application of Advanced Technologies for the

National Land Management" since 1999. In the

national land management (designated

hereinafter as NLM) framework, four important

tasks are needed to cope with the major missions

(Figure 1) including the safety against disaster. The research project includes applications of

recent advanced information technologies such as GIS, GPS and satellite images.

Among various missions included in NLM,

earthquake disaster prevention for civil

infrastructures is unique because;

1) Disaster mitigation requires various kinds of

information, i.e. hazard distribution, disaster

potential, and social activities,

2) Damage of facilities tend to be developed

simultaneously in wide area,

3) Urgent works are needed to prevent the expansion of disaster influence,

4) Disaster influence is unpredictable due to

the complexity of urban functions.

A seismic information system (designated

hereinafter as SIS) is one of the solutions to support disaster response activities (Figure 2).

Although advanced technologies have been

rapidly developed, their performance may not agree with the requirements of disaster response

activities. For example, satellite images may not be precise to detect damage of facilities, or

1) Head, Earthquake Disaster Prevention

Technology Division, PWRI, MOC

1 Asahi, Tsukuba, Ibaraki 3050804 Japan

2) Senior research engineer, ditto

63
Mission ^ Mission

Energetic

Activities Tasks

EfTicient Investigation

Observation X

Analyze

Pi aiming

Operation

Welfare

Safety against Disaster Mission

Fig.l Framework of NLM

Mission

Tasks Methods

ObservatioD x:

Wide Area

Monitoring

1 DaU

1 Collection

Earthquake Respons System

Analyze Analyze

Planoing

OperatktD

Decision Making

Personnel

SIS

Fig. 2 Earthquake Response System for NLM

maneuverability of the GIS is too complex for the emergency actions. Besides, other problems may

exist due to a framework of the SIS. For example,

sometimes the same kinds of data set are stored

to different computer systems redundantly. In addition, once a huge system is constructed, it needs much cost to extend or revise the system.

In other words, there are many possible problems

and losses depending on the SIS architecture.

Being aware of these issues, the PWRI has

composed a draft guideline for the efficient development of the SIS. In this paper, the brief contents of the "Draft Manual for Seismic

Information System Development" are introduced.

2. STRUCTURE OF MANUAL

2.1 Overview

For the efficient development of the SIS, the

Chapter 1: Introduction

Chapter 2: Basic Concept

Chapter 3 Requirements

Refer

Apply

Chapter 4 Element

Technologies

Chapter 5 System Safety

Fig.3 Structure of Manual

manual contains important topics, namely,

planning and development concepts of

user-oriented and flexible SIS, and a list of

state-of-the-art of technologies as a reference for the system planners. Also, the system safety issue

is included. The topics are arranged in corresponding chapters as shown in Figure 3. Note that the numbers of chapter are not of this paper but of the manual.

2.2 Key issues

Brief statements of the key topics in Figure 3 are

as follows. The detailed explanations are in the chapter three to six of this paper.

(1) Basic Concept of SIS Development

The basic concept of the SIS development

includes: concept for user-oriented SIS, concept

for flexible SIS, consideration of the system

interface, and the application of SIS in normal

state. These issues are fundamentals for the SIS

that satisfies requirements for earthquake disaster

response systems. While the concept mainly aims

to initial development of the SIS, they can be

applied to improve existing SIS with careful

consideration.

(2) Activities and Requirements for SIS

Classification of the earthquake response

activities, information required for the activities,

and functions required for the SIS are described

64
C

Bridge) Object

a Q

Fig.4 Data Sharing Architecture

Fig.5 Distributed Data Sharing Architecture

against earthquakes. Especially, SIS must be

flexible from the viewpoint of expandability

because it utilizes rapidly changing advanced

technologies. The better solution to cope with

this problem is a distributed data sharing system

(Figure 5). In such system, the data are stored in each server for specific function, and the data are

utilized by arbitrary clients through common

interfaces.

To realize the true distributed data sharing system,

the structure and the format of the data dealt by

SIS must be identified and clearly described. An

image of data structure that can include the

attributes and relationships is shown in Figure 6. Once the structure is defined and clarified, the information can be interpreted to any format. If

I Name ]

| Route "|

| Kilo post]

| Damage] AOribuies

Reference

Q Route 3 Earthquake

Terminal

Reference

Damage Image of Level Damage

Name Date of Occurrence

Fig.6 Data Structure

the format is well defined, arbitrarily added

subsystem can access the information regardless

with the types of operating systems and

applications.

4. REQUIREMENTS OF SIS

4.1 General

It is essential to investigate activities of the earthquake disaster response system and the information dealt with. In the manual, an example of such investigation in which response

activities are divided into four phases is shown.

For each phase, activities, information needed,

and functions required for SIS are clarified.

4.2 Earthquake disaster response phases

Assuming an earthquake disaster response for

civil infrastructures, response phases can be

classified as follows:

1) Emergency response phase: Overall rough

damage inspection is performed. Temporary

measures without heavy equipments are taken occasionally.

2) Temporary restoration phase: Activities

include damage inspection, decision-making

and activities for the temporary restoration.

3) Complete restoration phase: Similar to the temporary restoration phase, while it requires

65
in the chapter three of the manual.

(3) Review of Element Technologies for SIS

After requirements and structure of the SIS is determined, planners must investigate proper

technologies to realize the functions of the SIS. Because the SIS requires knowledge on both

advanced information technology and earthquake

engineering technology, planners need their comprehensive list. The manual provides an

inventory of current advanced technologies of

both fields in the chapter four.

(4) Consideration of Safety of SIS

Because the SIS is supposed to be used in an emergency situation, it must be well prepared for

a disastrous condition. Therefore, planners must

consider safety of devices, facilities and data

related to the SIS against earthquakes. Besides

the safety and security must be established and

maintained because the SIS is developed under

network situations. The technologies and

methodologies for the safety and the security are introduced in the last chapter of the manual.

3. BASIC CONCEPT

3.1 General

When one tries to develop an efficient SIS, there are several problems. First, advanced information

systems are usually developed mainly from a viewpoint of information technology

professionals who do not necessarily know the

activities of earthquake disaster response system.

As a result, sometimes personnel in the response

system encounter insufficiency of system

performances including the accuracy, speed and

capacity while some other performances exceed

their aims. Second, there can be another

problems related a system architecture. Once SIS

is developed as a large host-terminal system, it cannot be revised easily. Data may not be shared

easily between different systems that are separately developed, or even it may be difficult

to exchange data between different versions of

the same applications because of frequent

revision. In the first part of the manual, therefore,

several suggestions are stated to develop efficient and flexible SIS that meet with the requirements

of the earthquake disaster response activities.

3.2 Objective oriented concept

To develop the SIS that satisfies requirements of an earthquake disaster response system, SIS

should not be thought simply as an integration of

devices and application software. Rather, it is suggested that the system should be developed

based on the earthquake response activities and

the information dealt with in the activities. To

accomplish this objective, following three steps

are needed.

1) To investigate and organize disaster response

activities

2) To investigate information dealt with in the

activities

3) To clarify the requirements of functions to

efficiently deal with the information

Charts, tables and even the analytical method

such as the systems analysis are necessary to perform the tasks above.

3.3 Flexibility

SIS deals with many kinds of data of daily

maintenance in addition to the emergency

information after earthquakes. Therefore, it needs

to be able to access almost any information

regarding the facility management. One of the

solution of such data sharing system is servers with highly integrated functions and clients that browse the result of servers' functions (Figure 4), However, it has disadvantages such as

inflexibility, bottleneck and concentration of risk

66
more detailed inspection and complete

restoration of damaged facilities.

4) Response system support phase: This

phase is different from other three phases,

which concerns overall response activities from the start to the end. For example, it includes ensuring safety of the personnel

and management of resources to maintain

the response system.

4.3 Activities, information, requirements

for each phase

For each phase, response activities are itemized as shown in Table 1. For example the emergency response phase includes four

activities, namely, setup of earthquake

response system, emergency damage

inspection, investigation of response planning

and emergency response. Each activity

requires various kinds of information for its execution, and the functions of SIS to deal

with the information are listed.

The numbers in the brackets indicate the section numbers of the manual where the

technologies corresponding to the functions

are explained. Details of those technologies are explained in the chapter four of the manual.

The chapter three exemplifies that once the

planners analyzes the activities, information

and the required functions, they can

distinguish the appropriate technologies

applicable for SIS by referring the

corresponding sections.

5. ELEMENT TECHNOLOGIES

5.1 General

In the chapter four "Element Technologies for SIS" of the manual, numbers of technology

Table 1 Activities, Information and

Required Functions

Anivitio [nformation System F^uctions

Sel-upof EQ

R/spoose System

Earthquake Earthquake Detection(4J.l)

Facility Damage Damage Delection(4Ji4J-3)

Sel-up TrigRer Emergency Call(4.4.1)

Emergency Damage Inspection

Inspection Strategy Detection of Pha>omena(4J.MJ.5)

Inspection Activity FieW Activity Support(4J.4)

Facility Damage FiekJ Activity Support(4J.4) General Information Remote Sensine(4J5)

Study of Response Ptanning

Rdalfd Organ's Action Information Exchange(4.4J)

Construction Staffs [nformatioo Manaeement(4.4^)

fVioritv Decision Making Sunport(4.4.M.4J) Emergeocy ResDonse

Response Adivity FieW Activity Support(4J.4) Temporary Restoration Phase

Activities [nformatioo System Functions

Damage

Investigatioo

Investigation Strategy Remote Sensing(4

Investigation Activity FieW Activity Support(4J.4) Remote Sensing(4J.5)

Facility Damage

General Information

Restoration namiing

Related Organ's Action Information Ej(change(4.4J) Construction Stufls Infonnatioo Man2gement(4.4J)

Priority Decision Making Support(4.4.1-4.4J) Temporary

Restoration

Restoration Activity FieW Activity Support(4J.4) Complete Restoration Phase

Activities Information System Functions

Investigatioo

Investigation Strategy Decision Making SupportC4.6.M.6J) I Investigation Activity 1 FacilitY Damage HeW Activity Support(4J.4) General Information

Restoration

Related Organ's Action Information Exchange<4.4J)

Construction Stufls Information Managetiieot(4.4^) Planning Regional Restoration Plan Informalioa Exchange(4.4J)

Priority Decisioo Making Support(4.4.1-4.43) Complete

Restoration

Restoration Activity Fiekl Activity Support(4J.4)

Construction StuSs

Information Maiiageincnt(4.4.2) Information Exchan2e{4.4_3) ResDonse Svstem Supoort Phase

Activities Infomalion System Functions Response System

Related Information

Response System Level Common Infrastructure<42) Response System Staffs Information Management(4.4.2)

Disaster Prevention Base

Building Damage

Fiekl Activity Support(4J.4)

Device Damage

Activity Support

Stuffs Storage and Distribution Inforrttitioa Management(4.4.2)

42 43 Comnoo

rnfi!»rtructore 42.1

InformatioD Transfer

4.2.2 Information

Storing

Detcotioiiar .4-33

43.1 Earthquake

Detection

,43.2 Damage

Detection

Damage

Estimation 43.4

Field Activity Support

,43,5 Remote Sensing

4.4 4S 4.6 Commmlcatioa

4.4.1 Emergency

Call

4.4.2 Information Management

4.43

Information Exchange

DupUy

4.S.1 Map Display

4.S.2 Database

Operation 4.53

Still Image

Display

DedsMO Mskiog

Sopport

4.6.1 Route Search

4.6.2 Priority Optimization 4.63

4.S.4

Arrangement

Optimization

Moving Image

Display

Fig. 7 Element Technologies

that can be applied to SIS are introduced. The

67
planners of SIS can refer them to find the technologies suitable for the function of SIS that they have investigated in the chapter four. Also, it can be referred as the handbook to know the

state-of-the-art of the technologies for SIS. While

the advanced technologies are often explained

separately in respective specific area, this manual

collects them so as to be the comprehensive

reference for SIS planners. Also, it is intended

that the manual be useful for the personnel who

are not necessarily the professionals of the information technology to discuss about SIS with

the information technicians. The technologies

included in this chapter are divided into five

categories. The structures of this chapter with

subcategories are shown in Figure 7.

5.2 Common infrastructure technologies

This section introduces two kinds of fundamental

and important technology that are common to any

SIS. The information transfer technologies are

the vital for all kinds of information systems that includes wired/ wireless signal transfer technologies. In the manual, several kinds of

those technologies are listed and compared each

other as shown in Table 2. The Information storing technology section

focuses on the storage devices such as hard disk

drive and magnet-optic disk. Also, this section

introduces a brief concept of the relational

database that is related to the distributed data

sharing system. Fig. 8 shows an image of data

shared system with "Structured Query Language:

SQL" that is generally used to control RDB.

53 Technologies to detect phenomena

Prompt detection of phenomena caused by the earthquake disaster is necessary for the efficient response. This section includes following five

kinds of technologies.

Table 2 Information Transfer Technologies

Technologies Speed

Vulnerabilit)' Cost Service

Conflict Restrict|Dama^e Initial Run Area Telephone Line o X X i A oo OO OO

Ditto (ISDN) O X X A O OO O

Ditto (Packet) OO OO o A A A A

Cellular Phone O A A o OO O OO

Wireless (Ground) X OO oo 1 OO

Wireless (Satellite) X OO O i OO

Client PC

Fig. 8 Data Sharing using SQL

1) Earthquake detection: Technology to detect

the ground motion of earthquakes by

accelerographs.

2) Damage detection: Technology to directly

detect the damage of facilities by utilizing

various types of sensors.

3) Damage estimation: Technology to estimate

the damage of facilities indirectly based on

the monitored ground motion data.

4) Field inspection support: Combination of

technologies to support the field inspection of

damage.

5) Remote-sensing: Advanced remote sensing

technologies from platforms such as

helicopters, airplanes and artificial satellites. For example, sensors to detect damage should be

used properly depending on the distribution of

facilities and the required accuracy. While rough

detection of damage in the wide area needs only

simple sensors such as the wire sensor, more

sophisticated sensors are necessary for accurate damage detection for important facilities. Fig 9

illustrates an image of accurate damage detection

68
Fib^r Oplic Set^r

Wire Sensor

Fig.9 Damage Detection of Highway Bridge

Damage of Highway Bridges

Damage of River Levees

Fig. 10 Urgent Damage Estimation System

for a highway bridge.

Figure 10 shows an urgent damage estimation

system for highway bridges and river levees. The

system provides rough estimation based on the monitored ground motion data and pre-stored

structures database.

Field Inspection Activi

Positioning / Mobile

- G/'S JCommunication I

|f Operation/ Display

by Mobile Tools Damage State

I (Text, Code) Damage State

^ (SlilllMoving) /^

J Image

Damaged Site Acquisition

Fig. 11 Field Inspection Support Technologies

Field inspection support technology is illustrated

in Figure 11, in which various kinds of advanced

equipments are utilized. For example, personnel

of inspection team can easily understand their position in the damaged area with the help of the GPS and GIS devices. The information about the damage including texts, still and moving images

can be sent to the response center through the communication technology. Planners can develop

such a system with exclusive devices, or they can

use existing equipments in the market. Recent

technology can provide the flexibility of

development of the system.

Because of the extension of the earthquake

damage distribution, the information in wide area needs to be collected all at once. Remote-sensing

technology can be a candidate for such

requirement. The remote-sensing technology

consists of two parts, one is a platform and the

other is a sensor. Various kinds of platform are

characterized by their height, mobility and

stability. As for the sensor technology, it also has

various kinds such as an optical sensor, infrared sensor and laser profiler. They are characterized

by the ability to detect phenomena, accuracy and

69
Falling-off of Highway Girder

Still Image (Edge Strengthened)

Still Image (Colored)

Fig. 12 Damage Detection by Remote Sensing

influence of weather such as cloud and rain. Figure 12 shows simulated satellite images for understanding earthquake damage.

5.4 Technologies for information exchange

For the prompt and efficient earthquake response

activities, organizations and the personnel need

to exchange their information. This section

includes following three kinds of technologies:

1) Emergency call: It is vital to note the information for the establishment of the response system to persormel all at once.

Pager, voice modem and Internet technology

can be used.

2) Information Management: Information

collected and stored need to be appropriately

managed. For example, information of stock

materials for the restoration is meaningless if they are not updated timely. Information

Fig. 13 Data Exchange between Extranet

management issue is explained here although

it is rather a management technique than a pure hardware/ software technology.

3) Information exchange: Technology that is applicable to exchange information between

different organizations is introduced.

Figure 13 shows an image of the extranet

that utilizes the Internet technology. More

than one organizations exclusively exchange

their information through the Internet by

using extranet technology.

5.5 Technologies for display and analysis

While the information is gathered and stored, it becomes useful only when they are analyzed and

displayed in proper manner. This section includes

recent technologies to display maps, database and

still/ moving image.

Fig. 14 Still Image Displayed in GIS

70
While those technologies are not only for SIS but

general for any kinds of information system, their mechanisms are briefly discussed to help

planners grasp rough ideas to import them into

their SIS. For example, a still image linked with an object in GIS is shown in Figl4. In the manual

the mechanism of each technology is explained

conceptually.

5.6 Technologies to support decision-making

While the earthquake/ damage/ general

information itself improve the response activities,

it can be useful if the SIS provides more creative

information to support examination on efficient response alternatives. This section introduces

three kinds of examples:

1) Technology to find the shortest route

2) Technology to find priorities of emergency

restoration activities

3) Technology to find optimized pattern of

person/ material arrangement

Those technologies do not necessarily give the completely exact results, however, the earthquake

response team can investigate the appropriate

countermeasure by examining the alternatives.

6. SAFETY AND SECURITY

6.1 General

Vulnerability to disasters and security problems

are the negative aspects of SIS. It may be true if the system is not constructed and maintained

properly. This section introduces several measures for the reduction of these kinds of risks

to realize the functionalities of SIS both in the

usual and the emergency states.

6.2 Safety against earthquake disaster

SIS must be functional under the anticipated

level of earthquake. On the other hand, the information system devices are weak against

such conditions as shock, vibration and irregular

temperature during earthquakes. This section

explains following three kinds of issues to reduce

and manage the risk of SIS.

1) Constructing earthquake standing SIS: Many

points regarding the settings and

arrangements of system devices and facilities

2) Risk control with distributed SIS: Concepts

and techniques to keep the redundancy of the

system, network line and data

3) Preparedness for the critical situation: Importance of consideration of contingent

situation in which the response system must

act without sophisticated SIS

An Example of data distribution by using

Redundant Arrays of Inexpensive Disk: RAID

RAID 0 RAID 1

Fig. 15 Data Distribution using RAID

Reply to request from inside

firewall is permitted

unverified request is not permitted

Fig. 16 Security of SIS using Firewall

71
technology is shown in Figure 15. "RAID 0" is used to improve the disk access speed and

"RAID 1" is used to improve safety of data. There are other efficient and practical disk

system of RAID versions, and the planners can improve the data safety by utilizing them.

6.3 Security

Although the security does not directly related to the efficiency of SIS function, it is an important

issue regarding the computer network system.

Because the ideal SIS deals with almost all the

kinds of data of the earthquake disaster response

organization, the data stored and transferred in SIS must be well secured by such technologies as

firewall and encryption. Figure 16 illustrates an

idea of a firewall for a network system.

On the other hand, strong security occasionally

becomes an obstruction for disaster response

activities. An example is a case in which all the

limited authorized personnel are not accessible to the system in a disaster. Thus, the manual

suggests that both the security and its exception

from the point of the earthquake disaster

response need to be considered.

7. CONCLUDING REMARKS

This paper introduces the "Draft Manual for Seismic Information System Development" that PWRI recently developed. The manual was

prepared to develop efficient seismic information

system for civil infrastructure administrating

bodies. Highlights of the manual can be

summarized as follows:

1) Seismic information system is one of the

most sophisticated information systems to

' develop. Because of the characteristics of

earthquake disaster response activities.

user-oriented, feasible, earthquake standing

and highly secured system is required.

2) "User-oriented" seismic information system

must be studied based on the information

dealt with in each earthquake response phase.

Assuming the response activities for

infrastructure, disaster response phase can be

classified as emergency response, temporary

restoration, and complete restoration phase.

3) "Feasible" seismic information system must

be studied to achieve data sharing, risk

distribution, rational maintenance, and

extensibility of the system. This concept is essential not only for the construction of SIS

but also reconstruction/ upgrading occasions.

4) The manual describes element technologies

suitable for the damage detection,

communication, display and analysis, and

decision-making support. Technologies must

be carefully selected considering the cost and

required performance. It is essential to study

the performance including accuracy, speed

and capacity from the earthquake engineering

of viewpoints.

REFERENCES

1) PWRI: Guidelines for Disaster Information

Systems for Important Infra-structures, 25th

Joint Panel Meeting, UJNR, 1993.5

2) Sugita and Hamada: Development of Urgent

Damage Estimation System for Road

Facilities, 7th U.S.-Japan Lifeline Workshop,

UJNR, 1998.11

3) Sugita and Nozaki: Seismic Information

System for Civil Infrastructures, 30th Joint

Panel Meeting, UJNR, 1999.5

4) Earthquake Disaster Prevention Div.: Draft

Manual for Seismic Information System

Development, Technical Memorandum,

vol.3674, 1999.11

72
GIS and Pre- and Post-Disaster Risk Assessment

by

Stuart Nishenko', Claire Drury',

Scott Lawson^, Thomas Durham^, and Jawhar Bouabid^

ABSTRACT

In a pre-earthquake environment, loss estimates from credible scenario earthquakes

demonstrate the scope of the problem and

reinforce the continued need for structural and non-structural mitigation measures as a

central feature in long-term initiatives to reduce seismic risk in California and the

United States. Economic losses associated

with building damage from magnitude 7.0

scenario earthquakes along the Hayward

Fault in San Francisco are estimated to be

$37 billion (2000 dollars) and $80 billion

along the Newport-Englewood Fault in Los

Angeles. In both cases, building related

damage will contribute to between $6 and

$15 billion in losses due to business

interruption.

Loss estimates can also help determine the

significance of seismic design and

construction standards in reducing losses for

certain building types in earthquakes. This

information is key to understanding the role

of seismic rehabilitation in reducing

building losses from various levels of

ground shaking. Using the above scenario

examples, business interruption losses can

be reduced by >65%, economic losses to wood buildings can decrease by >33%, and

disruption to households and numbers of

displaced people can decrease by up to 95%

when the building stock is upgraded as part

of Comprehensive Rehabilitation Program.

The early collection, transmission and

mapping of earthquake ground motion data

has numerous scientific and emergency

management benefits. It provides vital information to emergency response

personnel to expedite field inspections,

provides quantitative data for damage and

loss estimation models, and prioritizes early

reconnaissance. Real-time data from the

National Earthquake Information Center and

TriNet were utilized by FEMA during the

October 16, 1999 Mw 7.2 Hector Mine CA

earthquake. Results from three simulations,

based on preliminary and revised epicenter

and magnitude data and TriNet /ShakeMap

data, demonstrate how damage estimates are improved as better data become available

with time.

KEYWORDS: Geographic Information

Systems, HazardsUS, HAZUS, scenario

earthquakes, seismic rehabilitation, TriNet,

ShakeMap

1.0 INTRODUCTION

This report summarizes studies by the

Federal Emergency Management Agency

(FEMA) to describe the economic and social

impact of scenario and actual earthquakes in

' Federal Emergency Management Agency

500 C Street SW, Washington DC 20472

^ Durham Technologies, Inc., 4200 Northside Parkway

Building 5, Atlanta, GA 30327

73
California. We also address the potential

reductions in direct economic and social

losses from the seismic rehabilitation of

buildings and the adoption of seismic design

standards.

These estimates were based on an

earthquake loss estimation methodology

HazardsUS (HAZUS) that was developed by

FEMA under a cooperative agreement with

the National Institute of Building Sciences

(NIBS). HAZUS was developed to provide

local, state, and federal officials in seismically vulnerable regions of the United

Sates with a uniform methodology for

estimating losses in future earthquakes

(Whitman et al., 1997; Jamieson and

Milheizler, 1997). HAZUS provides loss estimates for use by state, local, and regional

governments in planning for earthquake loss mitigation, emergency preparedness, and

response and recovery. The methodology is based on up-to-date ground shaking

projections from the United States Geological Survey (USGS) and the

California Division of Mines and Geology

(CDMG), extensive databases on the built

environment, and state-of-the art shaking to damage functions. FEMA is currently

engaged in developing similar loss estimation methodologies for wind and

flood hazards (Nishenko and Drury, 1999).

2.0 PRE-EARTHQUAKE STUDIES

2.1 Scenario Earthquakes

Policy, land-use and development decisions

are risk-driven and require appropriate

economic and social information to evaluate

different strategies and to assess the tradeoffs between costs and various benefits.

Loss estimates based on scenario

earthquakes provide a means to evaluate the

consequences of policy decisions.

2.1.1 San Francisco Bay Area

A recent United States Geological Survey

(USGS) study has identified the HaywardRogers Creek fault system as the most

hazardous in the San Francisco Bay area (USGS, 1999). FEMA (2000a) provides a summary of the losses from a magnitude 7.0

earthquake on the Hayward Fault for ten counties that touch the San Francisco Bay.

The major findings from this study include:

• Economic losses due to building

damage are estimated to be nearly $37

billion (2000 dollars), with over $16

billion occurring in Alameda County.

• Building damage will cause over $6

billion in losses due to business

interruption.

• When economic losses are characterized

by building type, it is estimated that wood buildings will account for over 50

percent of the total building losses.

2.1.2 Los Angeles

The Los Angeles Basin is underlain by at least a half a dozen faults capable of

generating earthquakes over magnitude 7.0. The Newport-Englewood fault was the site of the 1933 magnitude 6.3 Long Beach

earthquake and is considered to be one of

the greatest urban hazards in the nation.

FEMA (2000b) provides a summary of the

estimated losses from a magnitude 7.0

earthquake on the Newport-Englewood fault using HAZUS. An earthquake of this magnitude would potentially impact a six

county region in the Los Angeles Basin that has a population of over 15 million people.

Over 95 percent of the damages would occur

in Los Angeles and Orange counties, which

have the highest densities.

The major findings from this study include:

• Economic losses due to building

damage are estimated to be nearly $80

billion (2000 dollars). • 75 percent of the total economic losses

will occur in Los Angeles County.

• Building damage will cause over $15

billion in losses due to business

interruption.

74
2.2 Seismic Rehabilitation

In the San Francisco and Los Angeles

scenarios, the estimates of building damage

demonstrate the scope of the problem and

reinforce the continued need for structural and non-structural mitigation measures as a

central feature in long-term initiatives to reduce seismic risk. This section describes

how HAZUS was used to assess the impact

of seismic rehabilitation measures in reducing economic and social losses from

the Hayward and Newport-Englewood

scenario earthquakes.

The expected seismic performance of

buildings is determined by a combination of

parameters, including Building Design

Standards (low seismic design, moderate

seismic design, high seismic design) and

Building Construction Quality (code

compliant construction, non-code compliant

construction). By changing these

parameters, HAZUS can be used to test the

significance of seismic design and building

practices on the level of damage to buildings.

To determine the significance of seismic

rehabilitation in reducing losses from a magnitude 7.0 earthquake three 'buildingrelated' scenarios are used:

• Existing Condition - the building

inventory is assumed to be

representative of the existing building

composition in the San Francisco area

for the Hayward Fault scenario, and in the Los Angeles area for the NewportEnglewood Fault scenario. The

inventory includes a mixture of pre-code

structures with no seismic rehabilitation,

seismically rehabilitated structures, and

structures designed for various levels of

seismic resistance.

• Targeted Rehabilitation Program - the

most seismically vulnerable buildings

are assumed to be rehabilitated to meet a minimum level of seismic resistance

(protection of life safety).

• Comprehensive Rehabilitation Program - all buildings are assumed to be either

constructed^ or rehabilitated to the current seismic design standards of the

region.

For both the San Francisco and Los Angeles

scenarios,

• Business interruption losses could be

reduced by over 60 percent ($6.5 billion

to $2.3 billion in San Francisco and $15

billion to $5.7 billion in Los Angeles)

when the building stock is upgraded as

part of a Comprehensive Rehabilitation

Program.

• When wood buildings are upgraded as

part of a Comprehensive Rehabilitation

Program, economic losses decrease by

over 30 percent ($18.9 billion to $12.7

billion in San Francisco and $42.8

billion to $29 billion in Los Angeles)

• When the building stock is seismically

upgraded as part of a Targeted

Rehabilitation Program, major injuries

and deaths drop by 65 percent.

• Seismic rehabilitation has a significant

effect in reducing disruption to households caused by the earthquake.

Implementation of a Comprehensive

Rehabilitation Program could decrease

the number of displaced people by over 90 percent.

2.3 Estimated Losses to Building - By

Building Type

One of the objectives of these studies was to determine the significance of seismic design

and construction in reducing losses for certain building types in a scenario

magnitude 7.0 earthquake. This information

is key to understanding the role of seismic

rehabilitation in reducing building losses

from various levels of ground shaking.

Figures 1 through 4 illustrate direct

economic loss as a function of ground

motion for four basic building types-

• Residential- single family homewood frame construction

masonry construction

75
• Commercial - office building - low-rise

(< 4 stories)

concrete frame construction

masonry construction

The curves in Figures 1 through 4 were

developed for ground shaking from a magnitude 7.0 vertical strike slip

earthquake. Homes and commercial

structures are assumed to be located on deep

alluvium sites (NEHRP Type D soil) and are not subject to potential ground failure. Curves were developed for structures in the

Los Angeles area, and results are generally

applicable to the San Francisco Bay area as

well.

For each building type, three loss curves were developed that reflect three levels of

seismic design - No Seismic Design,

Moderate Seismic Design, and High Seismic

Design. For each building type, the loss curves show the estimated level of direct

economic loss (measured in dollars per

square foot) from ground shaking (measured

by the level of peak round acceleration

(PGA)). The direct economic losses include

the cost to repair the damaged structure, cost

to replace damaged contents, and temporary

living or business interruption expenses

incurred as a result of the earthquake.

All four building types dramatically

illustrate the benefits that are accrued by

moving from No Seismic Design to a Moderate Seismic Design standard - a

factor of two difference in direct economic

loss. Additional reductions in loss are realized by adopting High Seismic Design

Standards.

2.4 Limitations

Uncertainties are inherent in any type of loss estimation study. They arise in part from

incomplete scientific knowledge concerning

the severity and distribution of shaking from

earthquakes and their effect on the built

environment, and in part from the

approximations and simplifications

necessary for analysis of all facilities in a region. The range of uncertainty in HAZUS

is considered to be a factor of two to four. It

is natural to compare the results in this report with the output of similar models.

For such comparisons, the assumptions used

for each model need to be computable i.e., the exact location and magnitude of the

earthquake, the time of occurrence and the

time of occurrence, and the geographic area modeled are the same. The output should

represent damage and loss for the same

classes and inventories of buildings and

infrastructure.

3.0 POST-EARTHQUAKE STUDIES

In the immediate aftermath of a potentially

damaging earthquake, HAZUS can rapidly assess losses and guide emergency response

and recovery operations. HAZUS is currently being incorporated into FEMA's

Response and Recovery operations to support and facilitate the Information and

Planning mission of the Federal Response

Plan (FEMA, 1999).

The October 16, 1999 Mw 7.2 Hector Mine,

California earthquake provided an unique

opportunity to exercise the capability of

HAZUS in a real-time capacity.

Three scenarios were developed for loss estimates based on the availability of

information immediately following the

earthquake,

1. Preliminary epicenter and magnitude

(Ms 7.3) information from the National

Earthquake Information Center (NEIC)

along with a default rupture mechanism

generated by HAZUS and a default

building inventory scheme for the

impacted region.

2. A revised epicenter and magnitude (Ms

7.0) from NEIC and assuming the

rupture at the nearest fault (Bullion

Fault)

3. A TriNet-based ground shaking map

(ShakeMap, see Figure 5) along with an improved building inventory for the

impacted area.

76
Figures 6 and 7 clearly indicate how damage

estimates improved as better data became

available and were used. The number of

heavily damaged buildings was improved by

a factor of 15 while property damage losses were refined by a factor of six from the

initial estimate. Casualty and shelter

estimates were reduced by factors of 1 0 and

60, respectively.

4.0 CONCLUSIONS

The pre- and post-earthquake examples

discussed in this report demonstrate the

utility of the HAZUS loss estimation

methodology as a decision support tool.

In a pre-earthquake environment, loss estimates from credible scenario earthquakes

demonstrate the scope of the problem and

reinforce the continued need for structural and non-structural mitigation measures as a

central feature in long-term initiatives to reduce seismic risk in California and the

United States. These study results were

used to support the drafting of Federal

legislation that would provide incentives to encourage individuals, as well as state, local governments and businesses to invest in mitigation and earthquake loss reduction

('The Earthquake Loss Reduction Act of

2000').

In the immediate aftermath of a damaging

earthquake, HAZUS can rapidly assess

losses and guide emergency response and

recovery operations. Future activities to improve these capabilities include linking

HAZUS to other real-time networks in California as well as future USGS Advanced

National Seismic System (ANSS) stations

throughout the United States.

5.0 REFERENCES

FEMA, 1999, Federal Response Plan,

9230.1 -PL, Washington DC

FEMA, 2000a, Impact of a Magnitude 7.0 Earthquake on the Hayward Fault: Estimates

of Socio-Economic Losses Using HAZUS

(Hazards US),'FEMA, Washington DC

FEMA, 2000b, Impact of a Magnitude 7.0 Earthquake on the Newport-Englewood

Fault: Estimates of Socio-Economic Losses

Using HAZUS (Hazards US), FEMA,

Washington DC

FEMA, 2000c, Improving the Seismic

Performance of Existing Buildings:

Implications for Loss Reduction in the San

Francisco Bay area - a HAZUS (Hazards

US) Assessment, FEMA, Washington DC

FEMA, 2000d, Improving the Seismic

Performance of Existing Buildings:

Implications for Loss Reduction in the Los

Angeles Basin - a HAZUS (Hazards US)

Assessment, FEMA, Washington DC

Jamieson, G. and Milheizler, J., 1997, The

use of GIS in loss estimation and risk assessment, Proc. 29*^ Joint Meeting of the

US-Japan Panel on Wind and Seismic

Effects, UJNR, May 13-16, 1997, Tsukuba,

Japan.

Nishenko, S. and Drury, C, 1999, Recent

FEMA activities in earthquake risk analysis

and mitigation, EQ; Earthquake Quarterly,

Summer, 1999, 14-17.

USGS, 1999, Earthquake Probabilities in the

San Francisco Bay region: 2000-2030,

USGS Open File Rpt. 99-5 17

Whitman, R. V., Anagnos, T., Kircher, C,

Lagorio, H.J., Lawson, R. S., and Schneider,

P., 1997, Development of a national

earthquake loss estimation methodology.

Earthquake Spectra, 13, 643-662.

77
Figure 1 Potential damage to wood frame single-family residence from various levels of ground

shaking intensity for three levels of seismic design. The damage is quantified in terms of the

economic loss ($/ft^) and includes the cost to repair the damaged home, the cost to replace

damaged contents and any additional temporary living expenses caused by the earthquake.

60

Residential - Masonry - Earthquake Loss Curves

URM - Without Seiimic Design Considerations RM - Moderate Seismic Design Standards RM - lligh Seismic Design Standards

0.00 0.10 0.20 0.30 0.40 Peak Ground Acceleration (PGA)

(in g's)

O.SO

Figure 2. Potential damage to a masonry single-family residence from various levels of ground

shaking intensity for three levels of seismic design. The damage is quantified in terms of the

economic loss ($/ft^) and includes the cost to repair the damaged home, the cost to replace

damaged contents and any additional temporary living expenses caused by the earthquake.

78
Commercial - Concrete Frame - EQ Loss C^urves

0.00 0.05 0.10 0.15 0.20 0.25 0.30 0.35 0.40 0.45 0.50 Peak Ground Acceleration (PGA)

(in g's)

Figure 3. Potential damage to concrete frame office building from various levels of ground

shaking intensity for three levels of seismic design. The damage is quantified in terms of the

economic loss ($/ft^) and includes the cost to repair the damaged structure, the cost to replace

damaged contents and any additional business interruption expenses caused by the earthquake.

Commercial - Masonry - Earthquake Loss Curves

60 1 .. . . . .. , — :—-

. „.-; .— ,

. ,

0.00 0.10 0.20 0.30 0.40 0.50 Peak Ground Acceleration (PGA)

(in g's)

Figure 4. Potential damage to a commercial, low-rise masonry office building from various levels of ground shaking intensity for three levels of seismic design. The damage is quantified in terms

of the economic loss ($/ft^) and includes the cost to repair the damaged structure, the cost to replace damaged contents and any additional business interruption expenses caused by the

earthquake.

79
Pisgah-Bullion Mountain Fault

San Bernardino

Figure 5. Peak ground acceleration map for the October 16, 1999 Mw 7.2 Hector Mine

California earthquake based on real-time TriNet data.

80
1,000

800

S) 600

E

0)a.o

400

200

Scenario Scenario

Figure 6. HAZUS-based building damages from the three Mector Mine scenarios

500 800

Scenario Scenario

Figure 7. HAZUS-based social losses from the three Hector Mine scenarios

81
I
EARLY ESTIMATION SYSTEM FOR TSUNAMI HAZARD

KIRIYAMA Takaharu', OKAYAMA Kazuo2

ABSTRACT

The National Land Agency has created

Tsunami Hazard Maps of the entire coast of

Japan for the purpose of having a prior

estimate of the tsunami inundation range and

has developed an "Early Estimation System

for Tsunami Hazard" with the ability to use

the Tsunami Inudation Forecasting

Database(created from the results of the

calculations used to create the Tsunami

Hazard Maps) to estimate the tsunami

inundation range. The system searches the Tsunami Inundation Forecasting Database to find relevant locations based on the

forecasting zones announced in the new

tsunami forecasts of the Meteorological

Agency and the specific tsunami heights

expected. It then displays these locations on

the screen to provide a picture of the

tsunami inundation area and a rough

estimate of the number of buildings and

population contained there immediately after an earthquake.

KEY WORDS: Tsunami,

Hazard Map,

Database,

Tsunami Forecasting,

Early Estimation.

1. INTRODUCTION

Japan is surrounded by water and has suffered severe damage from the tsunami generated by

earthquakes, making measures to counteract

tsunami damage is an important issue for coastal areas. One effective means of

implementing tsunami countermeasures is to envision the tsunami that are likely to occur in a given region and create a "Tsunami Hazard Map"

that indicates, on a map, the areas where tsunami

inundation is expected.

Like earthquakes, when tsunami strike it is vital to have a rough estimate of the damage as quickly as possible. The tsunami that accompanied the Hokkaido Southwest Offshore

Earthquake of July 12, 1993 occurred at 22:17 but

virtually nothing was known about the extent of

the damage until the next morning because of the

fact that it occurred on the remote island of

Okushiri and because of the lateness of the hour.

Since fiscal 1996, the National Land Agency

has been operating an "Early Estimation System

for an Earthquake" that provides a seismic

intensities distribution map, building damage

estimate, and human damage estimate

immediately after an earthquake. Tsunami

estimation functions were recently added to this. This paper describes the creation of Tsunami

Hazard Maps of the entire coast of Japan for the

purpose of having a prior estimate of the tsunami

inundation range, and the development of the

"Early Estimation System for Tsunami Hazard"

with its ability to use a database (created from the

results of the calculations used to create the

Tsunami Hazard Maps) to estimate the tsunami

inundation range.

Dqjuty Director, Earthquake Disaster Countermeasures Division, Disaster Prevention Bureau, National Land Agency.

Director, Earthquake Disaster Countermeasures Division, Disaster Prevention Bureau, National Land Agency

83
2. CREATION OF TSUNAMI HAZARD

MAPS

2-1 OUTLINE

Tsunami Hazard Maps are maps that show the area that is expected to be inundated in the event of a tsunami attack. The map forms the basis for studies of tsunami countermeasures in the area. Figure 1 is an example of Tsunami Hazard Map

created by the National Land Agency. It shows at 1/25,000 scale the inundation area and inundation

depth for each meter when a tsunami of a certain

height attacks a certain stretch of coast. The map can be consulted to find the extent of

the tsunami inundation area, its depth, and the

streets, government institutions, public facilities, roadways, transportation facilities, factories and

the like located therein. This provides a clear picture of the tsunami disaster prevention tasks to be addressed in the area.

2-2 METHOD OF CREATION

In the past, the creation of Tsunami Hazard

Maps was to a large extent a matter of inference

based on past tsunami experiences. However,

experience may be inadequate for tsunami of

relatively low frequency.

The National Land Agency, in cooperation

with the Meteorological Agency and Fire and

Disaster Management Agency, has therefore

created a "Tsunami Damage Forecasting Manual"

that draws on recent research into tsunami and

recent advances in computer technology to formulate Tsunami Hazard Map creation methods

based on seismic fault models and tsunami

behavior simulations. The Tsunami Hazard Map

creation method of the National Land Agency

follows the method described in the Tsunami

Damage Forecasting Manual. An outline is provided in Figure 2.

(1) Set calculation range

The first step is to set the range of the initial calculation. We used rectangular areas of 30-50

kilometers per side for our ranges. These ranges were adjusted to conform with the forecasting

points used in the new tsunami forecasts of the

Meteorological Agency (described below). We set a total of 412 ranges covering the entire coastline

of Japan, including remote, populated islands.

We also set a seismic fault model tliat would

generate the largest conceivable tsunami for the range.

(2) Study past tsunami

We gathered information on local inundation

depths and seismic fault models etc. for representative tsunami of the past in order to verify the validity of the calculation results.

(3) Set calculation grid

We set the calculation grid at 100 meters. This

grid was chosen as a means of balancing precision

and costs.

(4) Digitalize topography

We digitalized topographic features using

topographical maps and ocean floor topographical

maps. For this, we considered only river areas of

200 meters or more in width and lakes with 200

meters or more contact with the ocean. We did not take into account the effect of levees or flood

gates.

(5) Set tsunami waveform

We set tsunami waveforms to produce tsunami

heights of 2, 4, 6, 8, and 10 meters in each area. However, for areas with a maximum tsunami

height of less than 10 meters in the seismic fault model, we did not consider tsunami heights above

the maximum.

(6) Set calculation conditions

We set the friction coefficient (ocean floor and land), calculation time interval, and other

conditions.

(7) Run numerical calculations

We ran numerical calculations using the

methods described in the Tsunami Damage

Forecasting Manual.

(8) Analyze results We analyzed results to determine if the

inundation areas and depths were reasonable

given the tsunami height and topography set. Where past tsunami examples were available, we

compared them against calculation results to judge the validity of results.

(9) Create Tsunami Hazard Map

We expressed calculation results on a

1/25,000 map. Calculation results were in the

form of data for a 100 meter grid, but we

84
expressed them as color-coded inundation depth

lines for 1 meter intervals.

(10) Create files containing calculation

results We created files containing the calculation

results for use as a database.

3. EARLY ESTIMATION SYSTEM FOR

TSUNAMI HAZARD

3-1 OUTLINE

The Early Estimation System for Tsunami

Hazard searches the Tsunami Inundation

Forecasting Database (created from the calculation results for the creation of the Tsunami

Hazard Maps) to fmd relevant locations based on

the forecasting zones announced in the new

tsunami forecast of Meteorological Agency and

the specific tsunami heights expected. It then

displays these locations on the screen. This gives

the government a picture of the tsunami

inundation area and a rough estimate of the

number of buildings and population contained

therein immediately after an earthquake (within

about 30 minutes). It is expected that this information will help to accelerate the government's initial response.

3-2 NEW TSUNAMI FORECASTS

The new tsunami forecasts were inaugurated

by the Meteorological Agency in fiscal 1999 and

use the forecasting zones shown in Figure 3 to display expected tsunami heights in concrete numerical terms (eight increments: 0.5 m, 1 m, 2

m, 3 m, 4 m, 6 m, 8 m, 10+ m).

The Meteorological Agency's model

envisioned approximately 4,000 epicenters in ocean waters up to about 600 kilometers off the coast of Japan and created a standard seismic

fault model for earthquakes occurring at these

points. The Agency then ran numerical

simulations of tsunami changing the magnitude

and fault depth of the earthquake to create a database of results (approximately 100,000

records). When an earthquake occurs, the system

searches the database of results for a corresponding epicenter and magnitude, and

displays the tsunami forecasts described above.

The Meteorological Agency database has

forecast points set every 20-30 kilometers along

the coastline. A single forecasting "^one will contain several focecast points; the tsunami height

displayed for the tsunami forecast is the maximum height for the zone.

3-3 SYSTEM FUNCTIONS

When a tsunami forecast is published, the Meteorological Agency sends the National Land

Agency on-line data on the tsunami height for each forecasting zone (in the future, this will also include tsunami heights for each forecast point).

The Early Estimation System for Tsunami Hazard

uses this data to search for corresponding

locations in the Tsunami Inundation Forecasting

Database and display them on-screen. The original Tsunami Inundation Forecasting

Database utilizes a 100 meter grid, but the system

converts this to 250 meter mesh data (using the maximum value for inundation depth in the mesh). This is done to simplify expression of

wide-range estimates. Expressing detailed

information on specific areas is not the purpose of

this system.

Below is a description of the information

displayed by the system.

(1) Wide-range estimates (Figure 4)

These displays are used to provide rough

pictures of wide-areas of 300 kilometers or more. This shows prefectural boundaries and the

maximum inundation depth on the coastline for municipal units.

(2) Middle-range estimates (Figure 5)

These displays are used to provide rough

pictures medium-range areas of 100-300

kilometers. They also display information in prefectural units. This shows municipal

boundaries and the maximum inundation depth on

the coastline for municipal units. The points and

figures in the ocean areas indicate the forecasting

point and the corresponding tsunami height.

(3) Narrow-range estimates (Figure 6, Figure 7) These displays show local conditions for areas of 100 kilometers or less, roughly in municipal

units. The map indicates municipal boundaries

and inundation depths at a 250 meter mesh.

(4) Record displays (Table I)

85
The system displays lists of potential

inundation areas in individual municipalities

(inundation depth of 2 meters or greater, and less than 2 meters), and of the number of buildings

and population included in these areas. The

building and population data is based on the three-dimensional mesh of the National Land

Numerical Data (approximately 1 kilometer

square), split up by number of land-area meshes

and converted to a 250 mesh.

The examples showing in Figures 4-7 and in Table 1 are models of the Nihonkai-Tyuubu

Earthquake that occurred on May 26, 1983. We

have estimated figures for forecast points from

past inundation records, and then input those

figures to search the Tsunami Inundation

Forecasting Database.

4. FUTURE TASKS

The Meteorological Agency began its new

tsunami forecasts in fiscal 1999. Luckily, Japan

has yet to experience a tsunami that would result

in the publication of tsunami forecasts, and

therefore the system has no track record. Nonetheless, the fact that this system can be used

to ascertain tsunami inundation levels directly

after an earthquake will have a significant impact

in accelerating initial government response

efforts. Currently, the Meteorological Agency

provides data only on tsunami heights for forecasting zones. It would be desirable in the near future to be given tsunami heights for individual forecast points in order to increase the

precision of estimates. It would also be desirable

to add functions to estimate the building

destruction and outflow, human damage, and

other specific damage from the tsunami.

BIBLIOGRAPHY

1) National Land Agency, Meteorological

Agency, Fire and Disaster Management

Agency, "Tsunami Damage Forecasting

Manual," (1997).

2) Meteorological Agency, "New Tsunami

Forecasts" (pamphlet).

3) Watanabe Hideo, Survey ofTsundfni

Damage in Japan (University of Tokyo

Press, 1985).

86
Table 1 Results of forecasting Tsunami damage

Prefecture Municipal Inundation

Area (2m —

)

Inundation

Area 2m)

Inundated

Buildings

Inundated

Population

All Japan 119.4km' 1072.6km' 224,941 403,831

Aomori Fukaura T. 11.4km' 7.5km' 2,012 3,900

Iwasaki V. 6.8km' 4.2km' 771 1,244

Akita 56.0km' 34.9km' 17,166 32,214

\A

Akita C. 7.0km' 4.6km' 1,876 5,751

Noshiro C. 9.5km' 6.2km' 2,557 5,236

OgaC. 16.5km' 11.3km' 4,810 9,745

Hatimori T. 5.2km' 0.6km' 1,411 1,919

Hachiryu T. 2.3km' 0.1km' 258 222

Minehama V. 2.3km' 0.3km' 75 219

AAA \AA/w

87
Figure 1 Tsunami Hazard Map (tsunami height: 10m)
® Set calculation range ^

Study

past

tsunami

Set calculation grid Set tsunami waveform

@ Digitalize topography > Set calculation

conditions

Yes

(D Create Tsunami Hazard Map ® Create files

Figure 2 An outline of creating Tsunami Hazard Map

89
Figure 3 Tsunami Forecasting Zones

90
Figure 4 Wide-range Estimates

Inudation Depth (m) — 10.0-

~~ 5.0- 10.0

2.0- B.O ~ 1.0- 2.0

0.&- 1.0

0.1 - 0.5
Figure 5 Middle-range Estimates

Inudation Depth (m) — 10.0

5.0- 10 0 2.0 - 5 0 - 1.0 u 0 o.&~ 1 0 0.1 - 0 6

92
Figure 6 Narrow-range Estimates ( 1

)

Inudation Depth (m)

H 10.0-

0- 10.0

b 0- 5.0

1 0 - 2.0

m 0 5 1.0

0 1 0.5

7

aomori-1

4.00 /

r- Fukaura T.

Hatimori T.

Minehama V. {

',

S

i 0 5 10 km

93
Figure 7 Narrow-range Estimates (2)

Inudation Depth (m)

10 .0

[. .0 ~ 10 0

I- 0 - 5 0

1 Ci - u 0 m 0 B - 1 0

0 1 0 5

I

Minehama V.

I

3S

4 km

_l

94
THEME 2

WIND ENGINEERING

95

RainAVind Induced Vibration of Bridge Cables in the United States

By

Harold R. Bosch'

ABSTRACT

The wind-induced vibration ot stay cables on

bridges in the United States has become a significant problem. Of the 24 cable-stayed

bridges that are currently in service, 10 have

exhibited some degree of cable sensitivity to wind. Several of these have suffered structural damage requiring repairs. This paper begins with

a very brief introduction to the problem of cable

response. Next, a review of cable-stayed bridges

in the U.S. is presented. This will be followed by

several examples of cable response problems and

retrofit measures to alleviate them. Finally, some

remarks regarding a national research study of

this problem and conclusions are provided.

KEYWORDS: aerodynamics; bridges; long- span, cable-stayed; cables; wind; rain.

1. INTRODUCTION

For more than 3 decades, engineers have been aware of the potential for wind-induced vibration

of cables on bridges [1-4]. In the 1970's, large

vibrations were observed on the Koehlbrand and

Brotonne Bridges. Mitigation measures were

devised and implemented at both sites. Many

other bridges around the world, especially cablestayed bridges, have exhibited similar cable

behavior in the years that have followed. With more than 600 cable-stayed bridges worldwide,

there are thousands of stay cables in service

representing a rather large inventory that has

potential for vibration problems.

There have been many investigations to identify

the origins of this problem and to develop various

methods of mitigation. While most of the

significant research has been conducted in Japan,

there have been notable studies in other countries

such as France and Germany. Despite what has

been learned about the problem, large amplitude

cable vibrations continue to be observed, even on

relatively new structures.

2. AERODYNAMICS OF CABLES

The interaction between wind and bridge cables

(stays) can result in vibrations that are often quite

large. Sometimes, the presence of rain is also

necessary for the vibrations to commence and the amplitudes to become significant. While research over the past 15 years has improved our general

understanding of the problem, the phenomenon

itself still is not fully understood.

Many factors can influence the response of bridge

cables to wind loading. These include, among

other things, wind speed and direction, cable

inclination, presence of rain, cable properties, and

structural damping. Response can be the result of

vortex shedding, galloping, rivulet formation,

buffeting, or aerodynamic excitation of the

roadway [5]. Both vertical and lateral vibrations can be significant, although the vertical motions

tend to predominate. Large amplitude motion has

frequently been observed under the following

conditions: low to moderate wind speed,

quartering yaw angle, light to moderate rain,

downward sloping cables (relative to wind),

moderate to long cable lengths, and low cable

damping.

Various mitigation measures have been

developed and implemented around the world.

(See Figure 1). Since structural damping is very

low in the cables, one approach is to supplement

it through the use of neoprene bushings, hydraulic

or viscous shear dampers, and cross ties. The

bushings serve a secondary purpose of

' Research Structural Engineer, FHWA TurnerFairbank Highway Research Center, 6300

Georgetown Pike, MeLean, Virginia 22101.

97
centering the cable in the guide pipe. The ties also serve to segment the cable span, which raises

the cable's natural frequency and critical wind

speed. The apparent elliptical shape of an

inclined cable presents some aerodynamic

stability problems, especially in the presence of

water rivulet(s). To address this, a number of

aerodynamic surface treatments have been used.

3. U.S. CABLE-STAYED BRIDGES

Despite their popularity elsewhere, cable-stayed

bridges did not begin to appear in the U.S. until the 1970's [6]. There are currently 24 of these

bridges that are in service. Nineteen of these are major highway bridges while the other 5 are low

volume or pedestrian structures. In addition,

there are another 6 bridges either in planning,

design, or under construction. The 30 bridge

sites are primarily located on our Pacific,

Atlantic, and Gulf coasts or along the Mississippi

and Ohio Rivers (Figure 2). The bridges are

steel, concrete, or steel/concrete construction

with single or dual towers and main spans

ranging from 33 to 396 meters. Cable types and

geometries vary with number of stays on each

bridge ranging from 12 to more than 200.

4. VIBRATION OF U.S. BRIDGE CABLES

For virtually all of the 24 cable-stayed bridges

currently in service, wind-induced response of the

cables was not directly considered during the

design. In addition, it has not been common

practice to consider this issue during the routine

wind tunnel testing and aerodynamic assessment

which generally takes place near the end of the

design phase. Formal guidelines for the aerodynamic design of cables do not currently

exist in the United States. It should be noted here

that most of the bridges were designed with guide

pipes and neoprene bushings, while the Sitka

Harbor and Luling Bridges do not have either. Only one bridge. Sunshine Skyway, was designed

with some special cable treatment and it has

hydraulic dampers. In contrast, most of the 6

bridges in design or under construction include some provisions to prevent large cable motions.

Approximately 10 of the in-service bridges have

exhibited some degree of cable sensitivity to wind (Table 1). Three of these (Dames Point,

Burlington, Hartman) were retrofit with cross ties during or shortly after construction. Cable ties on

the Fred Hartman Bridge began failing in Fall 1997 and all were removed in Spring 1998

(Figure 3). Most of the guide pipes have

fractured or cracked (Figure 4). Detailed

inspections and measurements have been

conducted on the Hartman and Cochrane bridges

and are underway on Veterans Memorial. A

redesigned restrainer system has been installed on Hartman as a temporary measure and prototype

dampers are being evaluated at both Fred

Hartman and Veterans Memorial sites. The

installation of hydraulic dampers and realignment

of guide pipes is planned at the Cochrane site. Measurements have been performed on the Weirton and Huntington bridges, prototype

dampers have been evaluated, and installation of

dampers is planned (Figure 5). Since cracks have

been discovered in three guide pipes on the

Talmadge Bridge, a detailed inspection and

evaluation is planned (Figure 6). Reports of large

cable vibrations on the Luling Bridge have been

infrequent, but significant nonetheless. Bridge

and cable behavior, as well as wind conditions at this site have been monitored on a long term basis

using remote instrumentation (Figure 7). This

monitoring will continue with greater focus on

cable activity and rain conditions.

5. NATIONAL RESEARCH UNDERWAY

In September 1999, the Federal Highway

Administration (FHWA) initiated a major

research project to advance our understanding of

this problem and to develop comprehensive

guidelines for the aerodynamic design of bridge

cables in the U.S. [7]. Twelve $tate Highway

Agencies (SHA's) have joined us as research

partners. (See Table 2). A contract has been

awarded to HNTB Corporation and work is well

underway on the information gathering phase.

The research team consists of HNTB staff as well

as key members from five other firms and several

prominent technical advisors. (See Tables 3 and

4).

98
6. CONCLUDING REMARKS

There is growing concern about the wind and

rain/wind induced vibration of bridge cables,

especially stay cables. Large amplitude motions

have been observed on a number of bridges in the U.S. and there has been structural damage

requiring repairs. The interaction of wind with

rain and cables is a very complex problem.

Although there have been many investigations in other countries, the cable behavior and the excitation mechanisms are still not fully

understood. A number of mitigation measures

have been developed which primarily involve

increasing the cable's structural damping or changing the cable's surface profile. Formal

design guidelines, which address this issue, do

not currently exist in the United States. Since

there is a need to advance our understanding of

this problem and to establish design guidelines, a major national study was launched in September

1999.

7. REFERENCES

[ 1 ] Langsoe, H E., Larsen, O.D., "Generating

Mechanisms for Cable Stay Oscillations

at the Faro Bridges", Proceedings of the International Conference on CableStayed Bridges, Bangkok, 1987.

[2] Hikami, Y., Shiraishi, N., "Rain-Wind

Induced Vibrations of Cables in Cable

Stayed Bridges", Journal of Wind

Engineering and Industrial

Aerodynamics, 1988,409-418.

[3] Matsumoto, M., et.al., "Rain-Wind

Induced Vibration of Cables of CableStayed Bridges", Journal of Wind

Engineering and Industrial

Aerodynamics, 1992, 201 1-2022.

[4] Ruscheweyh, H.P., "The Mechanism of

Rain-Wind-Induced Vibration", Wind

Engineering into the 21st Century,

Balkema, 1999, 1041-1048.

[5] Irwin, P.A., "Wind Vibrations of Cables on Cable-Stayed Bridges", Building to Last, ASCE, 1997, 383-387.

[6| Bosch, H.R., "Wind-Induced Vibration of

Bridge Cables: A National and

International Overview", Proceedings of

Workshop on Rain-Wind Vibrations in Cable Stayed Bridges, Atlanta, 1999.

[7] Bosch, H.R., "Wind-Induced Vibration of

Bridge Cables: Future Research - A

National Study", Proceedings of

Workshop on Rain-Wind Vibrations in Cable Stayed Bridges, Atlanta, 1999.

99
V Table 1. U.S. Bridges Exhibiting Cable Vibrations

Bridge When

Observed

Measurements Damage Action

WeirtonSteubenville

Construction Monitor and Evaluate

Prototype Spacer and

Visco-Elastic Damper

No Dampers Planned

Burlington Construction No No X-Ties Installed

Dames Point Construction INO INO X-Ties Installed

Fred Hartman Construction

In-Service

Monitor and Evaluate

Prototype Dampers

Significant Guide

Pipe Damage

(Detailed Inspection)

Original X-Ties

Failed

Temporary X-Ties

Installed

Cochrane In-Service Yes Minor

(Detailed Inspection)

Hydraulic Dampers

Planned

Realign Guide Pipes

Veterans In-Service In-Progress

Monitor and Evaluate

Prototype Dampers

No Monitor and

Evaluate

East Huntington In-Service Monitor and Evaluate

Prototype Dampers

No Hydraulic Dampers

Planned

Talmadge In-Service

(No Visual)

No Some Guide Pipe

Damage

Inspect and Evaluate

James River Construction General Measurements

During Construction

No None

Luling-Destrehan Construction

In-Service

Infrequent

Long Term Monitoring

Spans, Tower and Stays

No Continue to Monitor

100
Table 2. State Research Partners

State Cable-Stav Bridces

Virginia James River

Georgia Talmadge and Sidney Lanier

Kentucky Maysville and Owensboro

Florida Dames Point and Sunshine Skyway

Alabama Cochrane

Louisiana Luling-Destrehan

Texas Veterans Memorial and Fred Hartman

Mississippi US82

Illinois Quincy and Clark

Iowa Burlington

Missouri Cape Girardeau

South Carolina Charleston

Table 3. Key Team Members

Firm Area of Expertise

HNTB Corporation (HNTB) Project management, bridge design and analysis.

Johns Hopkins University (JHU) Flow induced vibration and wind engineering.

Buckland & Taylor (B&T) Bridge design, analysis and testing.

Rowan Williams Davies & Irwin (RWDI) Wind engineering, wind tunnel testing, damping

systems, dynamics and meteorology.

E*Sorb Systems (E*Sorb) Damping technology.

101
Table 4. Key Technical Advisors

Consultant Affiliation

R. Scanlan Johns Hopkins University

G. Fox HNTB

M. Matsumoto Kyoto University

0. Flamand CSTB Laboratory

Oil Dampers Viscous Shear Dampers

Araisv Brufge SaJcitama Bridge

(a) (b)

Axial Protaberances Cross Ties

Higashi-Keie BrUgt fnd Hartman Bridgt

(c) (d)

Figure 1. Common Mitigation Measures

102
I

Figure 2. Cable-Stayed Bridge Sites in U.S.

Figure 3. Damaged Cross Ties - Fred Hartman Bridge

103
wind I>Mmaee - Fmctured Cuidc Pipe Wind DamAge - Fractured Guide Pipe

Figure 5. Hydraulic Damper Detail - East Huntington Bridge

104
Wind Damage - Spalling and Crack

Figure 6. Guide Pipe Repairs - Talmadge Bridge

Cable Sensors

Figure 7. Cable Instrumentation - Luling Bridge

105

STUDY ON PERFORMANCE-BASED DESIGN METHODS FOR WIND EFFECTS ON

HIGHWAY BRIDGES

by

Hiroshi Sato^^ and Ken-ichi Ogi^^

ABSTRACT

Performance-based design methods for wind

effects on highway bridges are studied based on

the NKB level system. It was found that the existing wind resistant design methods for highway bridges in Japan could be easily

converted to the performance-based design

methods. To improve the design methods,

quantitative values for required performance, such

as a return period and allowable vibrational

amplitude should be studied further.

1. INTRODUCTION

In Japan, the wind-resistant design of

highway bridges up to the span length of 200m are conducted according to the Wind Resistant Design

Manual for Highway Bridges [1], [2] (hereinafter referred to as 'the Manual'), which was published

in 1991. This Manual is also applicable to the highway bridges up to the span length of 300m

with minor modification of its provisions. If span

length of a bridge becomes longer than 300m, the

specific design specification will be established

for the bridge. The basic concept of wind-resistant

design of such a long-span bridge is, however,

almost similar to that of the Manual.

Recently attentions have been paid to performance-based design methods. In the performance-based design methods, the performance level of structures is described

clearly, and designers have more freedoms in designing structures. In this paper, the existing wind resistant design methods for highway bridges and the NKB

level system are outlined first. Then the existing

wind resistant design methods are converted to the performance-based design format according to the NKB level system.

2. OUTLINE OF THE EXISTING WIND

RESISTANT DESIGN METHODS FOR

HIGHWAY BRIDGES IN JAPAN

The procedure of wind resistant design using

the Manual is as follows (Fig. 1). The wind

properties (basic wind speed, design wind speed,

turbulence intensity etc.) are decided first. The

design wind load is calculated, and the static design of the bridge is made.

At this stage, where major dimensions of

the bridge are determined, wind-induced

vibrations to be studied further are chosen

considering the design wind speed, the deck width

and the span length of the bridge. For a bridge

with short span length, no study on wind-induced

vibration is required.

For long-span bridges, wind-induced

vibrations specified above should be predicted

using the formulae provided in the Manual. Then

the wind-induced vibrations are evaluated. The

formulae were established considering past wind

tunnel test results. Since the formulae are expressed by means of only a few parameters, the prediction error of the formulae is not negligibly

small. Therefore, the formulae were established, so that their prediction may become safer. It means that evaluation 'good' based on the formulae is almost always 'good', however,

evaluation 'no good' based on the formulae is not always 'no good' because of the safety margin of

the formulae. In case that the evaluation based on

the formulae turns out to be 'no good', the bridge

engineer can modify the design or he can predict

wind-induced vibrations more accurately by means of wind tunnel testing. The evaluation

based on wind tunnel testing has priority over that based on the formulae.

1) Head, Structure Division, Structure and

bridge Department, Public Works Research

Institute, Ministry of Construction, Asahi 1, Tsukuba-shi, Ibaraki-ken, 305-0804, Japan

2) Research Engineer, ditto

107
3. PERFORMANCE STANDARDS AND

NKB LEVEL SYSTEM

3.1 Performance Standards

There are two kinds of design standards,

namely, prescriptive standards and performance

standards. The prescriptive standards describe the design and constituent materials of products [3]. On the other hand, the performance standards

define specific behavior of products. They define

requirements without imposing restrictions on the

form or materials of the solutions [3]. Therefore, a structural engineer can design

and construct the structure with more freedom

according to the performance standards. The latest and advanced technology can be applied to the

structure if it is verified to satisfy the requirements.

There may be a chance that more rational structures can be constructed according to the performance standards.

In the performance standards, however,

verification of compliance with the requirements

is more difficult than in the prescriptive standards.

It will take some time and money to verify the

performance of the structure or its elements.

3.2 NKB Level System

Nordiske komite for bygningsbestemmelser

(hereinafter referred to as 'NKB') developed what

is caUed NKB level system [4]. The NKB level system consists of five levels characterizing the purpose of building regulations. The five levels are illustrated in Fig 2. The levels are classified into three: Terformance requirements (Level 1 -

3)', 'Verification methods (Level 4)' and

'Examples of solutions (Level 5)'.

a) Deflnition of the levels The definition of each level [4] is shown

below.

Level 1: Overall objectives

The overall statement of the properties of a building that must be regarded as important

from the point of view of society and its individual members.

Level 2: Functional areas

The main properties specified as overall goal

level are classified in functional areas and

principles laid down for the realization of the specified intentions.

Level 3: Required performance

Operative requirements in order that principles

laid down under level 2 within the various

functional areas may be applied in the design

and construction of buildings.

Level 4: Verification

Instructions or guidelines laid down for verification of compliance with the requirements.

Level 5: Solutions

Supplement to the regulations with examples

of acceptable solutions, deemed to satisfy the regulations.

b) Examples of the levels The examples of contents of the levels can be

given as follows.

Level 1: Safety, serviceability, durability and so

on. Level 2: Safety of the structure or its element 'y'

to the effect 'z' caused by the action 'x'. Level 3: The intensity of the effect 'z' on the

structure or its element 'y' should be 'b' or less, when the intensity of the action 'x' is 'a'. Level 4: Analytical methods, testing methods,

calculation methods and so on. Level 5: If a structure that is designed and

constructed in accordance with some

specifications of prescriptive standards can be

expected to achieve the required performance, the

specifications can be also described here as examples of solutions.

c) Merits of the NKB level system

If structural design standards are described

according to the NKB level system, the

performance required for the structure will be

clarified in the Levels 1-3. Therefore the standards

have the merits of performance standards. The

purpose of each article of the standards will become clear. To overcome some demerits of the performance standards, namely difficulties in verification, the Level 5 can be utilized. Some of

108
the specifications of prescriptive standards that can be assumed to satisfy the required

performance can be included in the Level 5.

4. PERFORMANCE-BASED DESIGN

METODS FOR WIND EFFECTS ON

HIGHWAY BRIDGES

The wind resistant design methods described

in the Manual are converted to the performance-based design format according to the NKB level system.

4.1 Overall Objectives (Level 1) A highway bridge shall be planned and

constructed in such a way that its safety and

serviceability be obtained and maintained under

the wind actions that are expected at the bridge

site during the service life of the bridge.

4.2 Functional Areas (Level 2)

Wind actions that shall be considered in the design of a highway bridge are as follows:

a) wind load; b) gust response;

c) flutter; d) galloping;

e) vortex-induced vibration.

The wind actions may produce the following

effects:

f) excessive forces, stress or deformation of the bridge or its elements;

g) repeated dynamic stress or deformation

causing fatigue of the bridge or its elements;

h) aeroelastic instability;

i) excessive dynamic movements causing concern or discomfort to users of the bridge.

43 Required Performance (Level 3)

a) Wind load and gust response

For the wind load and gust response, the

safety of the bridge shall be examined in viewpoint of the wind effects referred to in 4.2 f). The design wind load Pd for static design

may be obtained from the following formula.

Pd = 0.5 P Ud" Cd An G (1)

where, p : air density, Ud: design wind speed, Cd: drag coefficient, An: projected area, G: gust factor. The resultant forces, stress or deformation

shall be less than the allowable values specified in the Specification for Highway Bridges.

b) Flutter and galloping For flutter and galloping, the safety of the bridge shall be examined in viewpoint of the wind

effects referred to in 4.2 h). The following inequalities shall be satisfied,

Ucf > U, — Flutter (2)

Ucg > U, — Galloping (3)

U, = Ud E,i E,2 (4)

where, Ucf: critical wind speed for flutter, Ucg:

critical wind speed for galloping, U,: reference

wind speed for flutter and galloping, Ud: design

wind speed, Eri: correction factor for the effect of

gust, Er2: safety factor(1.2).

The design wind speed for dynamic design

Ud may be obtained from the following formula.

Ud = UioEi (5)

where, Uio: basic wind speed, Ej: correction factor

for altitude and terrains. The basic wind speed Uio is defined as the mean wind speed over open farmland at an

elevation of 10m, averaged over a period of 10

minutes. Its return period is 100-years. The

probability that the annual maximum mean wind

speed exceeds the basic wind speed is 40% in a 50-year period. The basic wind speeds in Japan were classified into 4 categories, namely 30m/s,

35m/s, 40m/s and 45m/s.

c) Vortex-induced vibration

For vortex-induced vibrations, the safety of

the bridge shall be examined in viewpoint of the wind effects referred to in 4.2 f) and g). The

serviceability of the bridge shall be examined in viewpoint of the wind effects referred to in 4.2 i). Unless the following inequality (6) is satisfied, the maximum amplitude of the vortex-induced vibration shall be less than the

109
allowable amplitude shown in equations (7) and

(8).

(6)

where, wind speed for the maximum

amplitude of vortex-induced vibration.

h3=0.04/fh — bending (in m) (7)

6 3=2.28/(b ie ) — torsion (in deg.) (8)

where, h^: allowable amplitude for bending

vortex-induced vibration, fh: natural frequency of

the 1st bending mode, 6 ^: allowable amplitude

for torsional vortex-induced vibration, b: distance

between the deck center and the center of outmost

lane, f^: natural frequency of the 1st torsional

mode.

4.4 Verifications (Level 4)

The wind-induced vibrations such as flutter, galloping and vortex-induced vibration may be

predicted either by the formulae or by wind tunnel

tests. The predicted critical wind speed or maximum amplitude shall be compared with the reference wind speed or the allowable amplitude

described in 4.3.

a) Prediction of wind-induced vibrations by

the formulae

The formulae to predict wind-induced

vibrations are provided in the Manual. They were

established by considering past wind turmel test

results. The formula for prediction of critical wind

speed of flutter is as follows:

Ucf = 2.5 f^B (9)

where, Ucf: critical v^nd speed for flutter, f^:

natural frequency of the 1st torsional mode, B: deck width.

Galloping may take place for solid steel deck

whose deck width to depth ratio (B/d) is less than

5, and when turbulence intensity is small. The

formula for prediction of critical wind speed of

galloping is as follows:

Ucg = 8 fh B ; when angle of attack is almost 0

deg. (10)

Ucg = 4 fh B ; when angle of attack is positive.

(11)

where, Ucg: critical wind speed for galloping, fh: natural frequency of the 1st bending mode.

Vortex-induced vibrations may take place for solid deck. The formulae for prediction of critical wind speed and amplitude of vortex-induced

vibrations are as follows:

Ucv = 2.0 fh B — bending (12)

Ucv= 1.33 fe B — torsion (13)

Ac = En^ E, (14)

A, = )3ds- 0.05- (d/B)/(m, 6^) — bending (in h/B) (15)

A, = )3d3- 13.2- (d/B)'/(Ip, (5 e) — torsion ( in deg.) (16)

where, Ucv: wind speed for the maximum

amplitude of vortex-induced vibration, Ac:

corrected maximum amplitude of vortex-induced

vibration, A^: maximum amplitude of

vortex-induced vibration for rigid model in smooth flow, En^: correction factor for vibrational

mode (about 4/7i:), E,: correction factor for the

effect of turbulence, iS^^: correction factor for sectional shape, m,, Iprt reduced mass or reduced mass moment of inertia, (5 h, ^ s : structural damping (logarithmic decrement).

b) Prediction of wind induced vibrations by

wind tunnel tests Wind tunnel tests increase the reliability of

wind resistant design of bridges. Since the prediction of wind-induced vibrations based on

the formulae may often provide safer values, there

is a fair chance that wind tunnel study provides more economical design.

Wind tunnel testing methods applicable to the prediction of wind-induced vibration of

highway bridges are as follows: full aeroelastic

model test, taut-strip model test, and

spring-mounted rigid model test.

4.5 Solutions (Level 5)

According to the types of a bridge, main span

length L, deck width B and design wind speed Ud,

the bridge may be assumed to satisfy the required

110
performance for some wind-induced vibrations.

The conditions that a bridge can be assumed to satisfy the required performance for some

wind-induced vibrations are shown in Table 1.

4.6 Discussions

The return period for design wind speed is the most important and fundamental value for wind resistant design. The rational return period

for design wind speed should be studied further by

comparing with the return periods for other

natural disasters, such as earthquakes.

The allowable amplitude for the vortex-induced vibration shown in equations (7)

and (8) were determined by considering full-scale data of the vortex-induced vibration and bridges

that countermeasures were actually applied to. The rational allowable amplitude should be

studied further, especially in viewpoint of

serviceability.

5. CONCLUSION

The performance-based design methods for wind effects on highway bridges were studied.

The wind resistant design methods described in the Wind Resistant Design Manual for Highway

Bridges were converted to the performance-based

design format according to the NKB level system.

It was found that the existing wind resistant design methods could be easily converted to the

performance-based design methods, because the required performance and verifications are clearly

described in the existing methods. Some of the quantitative values for required performance, such

as a return period for design wind speed and

allowable vibrational amplitude, however, should

be studied further to improve the design methods.

REFERENCES

[1] K. Yokoyama and H. Sato: On the proposed

wind resistant design manual for highway bridges

in Japan, International Colloquium on Bluff Body

Aerodynamics and its Applications, 1988, 10,

(Journal of Wind Engineering and Industrial Aerodynamics, 33, 1990, pp.395-405, Elsevier

Science Publishers)

[2] H. Sato and T. Mabuchi: Development of

manual and expert system for wind resistant design of highway bridges, Transaction of the Jubileum Conference on Wind Effects on

Buildings and Structures, Part II

, 1998, 5, pp. 69-76

[3] ISO: Performance standards in building —

Principles for their preparation and factors to be

considered , IS06241- 1984 (E) , 1984

[4] Den nordiske komite for bygningsbestemmelser: Strokturering av Byggregler (Structure for Building Regulations),

NKB report No 34, 1978

111
Tablel Bridges that can be assumed to satisfy the required performance for the wind actions

Type of girder Wind actions Truss girder for cable-stayed or suspension bridges

can be assumed to satisfy the required performance for;

flutter, if L Ud/B<350;

galloping;

vortex-induced vibration.

Open solid girder for cable-stayed or suspension bridgescan be assumed to satisfy the required performance for:

flutter, if L Ud/B<350;

galloping, if L Ud/B<330, B/d>5, Iu>0.15 or if it is made of concrete; vortex-induced vibration, if L Ud/B<200 or if Iu>0.20.

Closed solid girder for cable-stayed or suspension bridges

can be assumed to satisfy the required performance for:

flutter, if L Ud/B<520;

galloping, if L Ud/B<330, B/d>5, Iu>0.15 or if it is made of concrete; vortex-induced vibration, if L Ud/B<200 or if Iu>0.20.

Steel girder bridges can be assumed to satisfy the required performance for:

flutter; galloping, if L Ud/B<330, B/d>5 or if Iu>0.15;

vortex-induced vibration, if L Ud/B<200 or if Iu>0.20.

Bridges other than the above typescan be assumed to satisfy the required performance for:

flutter; galloping;

vortex-induced vibration.

112
START

WIND PROPERTIES

3^

DESIGN WIND LOAD

STATIC DESIGN OF THE BRIDGE

NO SHOULD WIND-INDUCED

VIBRATION BE STUDIED?

YES~

PREDICTION OF WIND-INDUCED

VIBRATION BY FORMULAE

GOOD

DESIGN - MODIFICATION

EVALUATION OF WINDINDUCED VIBRATION

NOT GOOD

DESIGN MODIFICATION OR

MORE PRECISE PREDICTION?

<MORE PRECISE PREDICTION

PRECISE PREDICTION

MINOR MODIFICATION

OR MAJOR?

MINOR

MAJOR

DESIGN TOWERS AND

BRIDGE MEMBER

DESIGN AT THE

ERECTION STAGE

END

Fig. 1 Procedure of wind resistant design using the Manual

113

WEATHER SITUATION ON GROUND OBSERVEDAT FIRE STATIONS

- CASE OF TYPHOON VICKI

by

Yasuo 0KUDA'\ Hirohiko ISHIKAWA^', Takeshi FUJII^^ and Hisashi OKADA^'

ABSTRACT

Collecting a weather observation data at fire stations located in Kinki district and Chubu

district, the weather situation on the ground of

typhoon Vicki was investigated. There were four severe storm areas where the instantaneous wind

velocity exceeds 50m/s in Kinki district. The

severe storms occurred in the two cases: at the

almost same time when the daily minimum

atmospheric pressures appeared and at about one

hour after the daily minimum atmospheric

pressures appeared. Comparing the severe storm

areas with the combined radar echo chart, it was showed that the severe storms of the latter case occurred around the rain band to form in the rear of the typhoon and the direction of the severe storms was divided into the two directions;

tangent and normal directions against the

rotation of the typhoon. The detailed weather

situation on the groimd of the typhoon became

clear about temperature, humidity, precipitation,

and atmospheric pressure. In this way, the

weather observation data at the fire stations was

showed very useful in order to collect weather

information on the ground of the typhoon.

KEY WORDS: Typhoon Vicki

Fire Station

Weather Situation

Combined Radar Echo

1. INTRODUCTION

Typhoon Vicki was landed at about 1 p.m.

September 22nd, 1998 at the northern region of

Gobo city in Wakayama prefecture, following

typhoon Waldo that was landed in Wakayama

prefecture previous day. Typhoon Vicki ran up

through Kinki district and exerted enormous

damages on each place in Kinki district, mainly

in the Nara prefecture. As this typhoon passed

the central region of Kinki district as well as the

typhoon 7916, the weather elements of the

typhoon on the ground was observed in a lot of

places at the fire stations and the schools in addition to the meteorological observatories and

so on.

In recent years, the number of the fire fighting

headquarters is increasing where the weather

data acquisition systems for the fire fighting were installed in the local governments in the

whole country. The weather data acquisition

system for the fire fighting observes the

weather elements

1 ) Head, Department of Structural Engineering,

Building Research Institute, Ministry of

Construction

2) Associate Professor, Disaster Prevention

Research Institute, Kyoto University

3) Professor, General Education and Research

Center, Kyoto Sangyo University

4) Director, Department of Structural

Engineering, Building Research Institute,

Ministry of Construction

automatically for the prevention against disasters

and the publicity work in the local area. There

are tens fire fighting headquarters per each

prefecture in Kinki district and Chubu district. The weather data acquisition system for the fire fighting is installed at the most of the fii-e fighting headquarters. Therefore, the weather

observation network with this fire fighting

headquarters becomes much closer than the

weather observation network of the

meteorological observatories that are only one or two places per prefecture. In the great city, the

weather observation network with this fire fighting headquarters becomes closer than

115
AMeDAS (Automatically Meteorological Data

Acquisition System, four measured elements of

precipitation, wind vector, sunshine and

temperature). Moreover, this weather data

acquisition system can measure the weather

elements in a time series, which are indispensable to observe typhoons such as the

instantaneous wind velocity, the atmospheric

pressure, and the humidity, which were not measured in AMeDAS. This weather data

acquisition system is expected to be ideal for the

investigation of the ground weather elements of

the typhoon.

Uematsu et al.(1999) made a data base about the

weather observation situation at the fitt stations

in Aomori and Akita prefecture in Tohoku

district. However, there was no report to verify

the utility of this weather observation network

by collecting the weather data observed with

such weather data acquisition systems from the

extensive area. Because the purpose of this weather data acquisition system itself seems to provide the weather information only for the purpose of the area.

Table 1 Number of fire stations

Name of prefecture Number of fire station Number of answer Rate of answer (%)

Kinki district Kyoto

Mie

Shiga

, Osaka

Wakayama

Nara

Chubu district Aichi

Gifu

Fukui

17

16

11 33

21

13

47

24

12

10

11 7 64

22 67

8

11 85

38

13 54

8 67

59

69

38

81

Total 194 128 66

Table 2 Total Number of measuring points

Classification of measuring points Number

Fire station 128

Meteorological observatory 27

AMeDAS(four measured elements) 71

Power station 27

Total 253

116
Fig. 1 (a) Path of Typhoon Vicki

Fig. 1(b) Weather chart at 9:00 on 22nd September.

Therefore, the authors requested the provision of

the weather data after the passage in typhoon Vicki

in the purpose to examine the ground weather

situation in typhoon Vicki in detail. As a result, a

lot of answers were gotten as shown in Table 1. Especially, the answer percentage of Nara and

Aichi prefecture exceeded 80%. After that, it gets

a similar weather data from the weather

government service station (Meteorological

Agency, 1998a) and the power stations in Kinki

district. The weather data could be gotten from

253 observation points in all as shown in Table 2. Here, it reports the result examined about the time

series change of the weather elements with the

typhoon passage charts, compared with the rainfall strength figtires from the combined radar echo

charts Based on this result, it verifies the utility of

the weather observation netv,'ork observed at the

fire stations.

2. WEATHER GENERAL CONDITION OF

TYPHOON VICKI

The course in typhoon Vicki (Meteorological

Agency, 1998b) and the weather chart at 9:00 on

September 22nd (Meteorological Agency, 1998b)

are shown in Figs.l(a)(b). The weak tropical

depression, which occurred on the western sea of

Philippine Luzon Island at 15:00 on September

17th, developed into typhoon Vicki at 21:00 on the same day. It developed gradually on the course which is shown in Fig. 1(a) and hit Kii peninsula of

Japan around 13:00 on September 22nd. The

minimum atmospheric pressure at the center of

typhoon Vicki was 960hPa at the landing. The

maximum wind velocity was 35m/s around the

center. The radiuses of the severe storm area where the wind velocity was over 25m/s were 190km at the southeast side and 150km at the

northwest side. The radiuses of the severe storm

area where the wind velocity was over 15m/s were 700km at the southeast side and 220km at the

northwest side.

After the landing, it ran from the southwest to the

northeast in Kinki district while weakening

gradually the power of the typhoon and went

through to Toyama bay. Moreover, it re-landed on Yamagata prefecture before 22:00 on the same

day. The central atmospheric pressure at the

117
Fig.2 Location of fire stations meteorological

observatories and power stations

re-landing was 990hPa and the maximum wind

velocity was 25m/s near the center. The radiuses

of the severe storm area where the wind velocity was over 15m/s were 460km at the southeast side

and 190km at tiie northwest side. Then, it went

through to the Pacific Ocean and it became the

extra tropical depression at 9:00 on 23rd

September. This typhoon was a very ordinary

typhoon at size and strength but this typhoon

moved very fast and accompanied the heavy

rainfall in the influence of the front. The

movement speed (Meteorological Agency, 1998b)

of this typhoon is about 50-55 km/hr immediately

before Kii peninsula landing and after the landing

it is raising speed in about 75-85 km/hr.

3. WEATHER SITUATION ON GROUND

3.1 Weather Observation Point

The position of the fire fighting headquarters, the

weather government service station and the

Fig.3 Path of Typhoon 9807(Vicki) on Kinki

district

power stations, where the observation data could

be got, shows in Fig.2. AMeDAS is equally

arranged approximately over all the prefectures,

but one or two meteorological observatories were

located per prefecture. On the other hand, the

observation points of the fire fighting headquarters concentrate on the urban regions of the great city

such as the Osaka or Nagoya city. There are few

numbers of the observation points of the fire fighting headquarters in the ravine regions.

However, a lot of observation points at the power

stations exist in the ravine regions and the coastal

region in Kinki district. They complement an

observation network in the fire fighting. In the

northern region of Nara prefecture where the large

damages were sustained in the typhoon Vlcki, the

number of the meteorological observatory was

only one and the number of AMeDAS was only

three. But the number of the fire fighting

headquarters was 11 and more detailed weather

information on the ground could be collected. This

figure shows that the weather observation network

at the fire stations formed the close observation

network mainly in the urban region is very

effective in order to grasp the damage situation of

the severe storm disaster on the urban region like this typhoon.

118
3.2 Course of Typhoon Vicki

The wind directions at observation points changed

to be clockwise at filled circle marks and

anticlockwise at empty circle marks shown in Fig.3. The boimdary that approximately divided

Kinki district into east and west certainly existed.

The thick broken line showed this boundary. This

line is equivalent to the course in the wind center of the typhoon. As the times when the wind

direction changed were recorded at some

autograph observation points, the position of the

wind center of the typhoon was shown by every

hour on Fig.3. Also, the position of the

atmospheric pressure center of the typhoon which was looked for every 10 minutes by the objective

analysis(Fujii et al., 1999) was shown by double

circle mark. Fujii et al. used atmospheric pressure

records by the weather government service. The

triangle mark is the position that the

Meteorological Agency mainly looked for every

hour. It finds that each course of the typhoon was

meandering and that the course in the wind center and the atmospheric pressure center of the typhoon

resembled very well each other but that the course

in the wind center shifted to the direction of the

progress than the course in the atmospheric

pressure center with the about 10 to 20km left.

33 Situation of Severe Storm

The distributions of the average wind velocity

value for 10 minutes (the average fi^om 50' to 00'

in every hour) and wind velocity vectors are shown by the time series fi-om 13:00 to 16:00 in Figs.4(a)-(d). In the figure circle marks show the

position of the atmospheric pressure center of the

typhoon that was found at each time fi^om the

objective analysis (Fujii et al., 1999). In 14:00,

when the atmospheric pressure center of the

119
Osaka and Nara prefecture, the severe storm area where the average wind velocity exceeds 20m/s

appears in the rear of the typhoon in addition to the right side of the typhoon. In Wakayama city,

the maximum wind velocity recorded 32.4m/s at 14:20. After that, this severe storm area passed

Osaka, Nara and Mie prefecture and reached to the

western region of Aichi prefecture and the

southwest region of Gifii prefecture. The severe storms were recorded over 30m/s respectively, i.e.; maximum wind velocity 31.9m/s at 15:48 in Kameyama city in Mie prefecture, the same

32.7m/s at 15:57 in Kaitsu town in Gifti prefecture. As shown in the wind velocity vector of Fig.4, the wind direction in the severe storm

area formed in the right of the typhoon was mainly

the southeast to the south, but the wind direction

in the severe storm area in the rear of the typhoon was mainly south to the west.

Fig.5 Distribution of daily maximum

instantaneous wind velocity (m/s)

Fig.5 is the distribution of the daily maximum

instantaneous wind velocity in each observation

point. In Fig.5 the circle mark shows the position

of the atmospheric pressure center of the typhoon

looked for from the objective analysis (Fujii et al., 1999). The severe storm area, where the maximum

instantaneous wind velocity exceeds 40m/s, spread

from the southern region of Kinki district to the

western region of Chubu district. The daily

maximum instantaneous wind velocities were

recorded 51.0m/s in Tanabe city at 13:13, 50.0m/s

in Wakayama city at 14:13, 50.8m/s in Sumoto

city at 14:07, 59.5m/s in Shinjo town at 14:59 and

56.4m/s in Ueno city at 15:14.

-dll.flO. Fig.6 Difference hours between occurrence time of

daily maximum instantaneous wind velocity and

daily minimum atmospheric pressure

Fig. 6 shows the difference of occurrence times

when the daily maximum instantaneous wind

velocity and the daily minimum atmospheric

pressure were recorded at each observation point.

The figure is divided into the two areas; the

differences of occurrence times were almost zero (A in the figure) and more than one hour(B in the

figure). In Tanabe city the daily maximum

instantaneous wind velocity and the daily

minimimi atmospheric pressure were recorded at the almost same time, but in Sumoto city,

Wakayama city, Shinjo town and Ueno city, the

daily maximum instantaneous wind velocity was

120
(a) 13:00

(b)14:00

(c) 15:00

atmospheric pressure occurred.

3.4 Situation of Temperature

The distributions of the temperature from 13:00 to 16:00 are shown in Figs.7(a)-(d). The absolute

value of temperature can not be compared each

other because it depended on the altitude of each

observation place. The distribution of the

temperature in Fig. 7 is showed in the difference

from the day average value at each observation

point. In the figure shows the position of the atmospheric pressure center of the typhoon at the

time that was foimd from the objective analysis

(Fujii et al, 1999).

The cold area spread on the western side around

the center of the typhoon when the typhoon hit. From 14:00 to 15:00, this cold area was distributed

like a hook from the western side around the center of the typhoon. It can be understood that the

cold air entered from the northwestem side into

the center of the typhoon. At this time the interval

of the isotherm became denser from the southern

region of Osaka prefecture to the northern region

of Nara prefecture and the temperature inclination

in space became very large. In this way, the cold

air from the northwestem side of the typhoon

found to be brought the rapid decline of the

temperature across the border between Osaka and

Nara prefecture.

3.5 Situation of Precipitation and Specific

Humidity

The distributions of the hourly precipitation from

13:00 to 16:00 are shown in Figs.8(a)-(d). In the

figure shows the position of the atmospheric

pressure center of the typhoon at the time that was

found from the objective analysis (Fujii et al., 1999). As typhoon Vicki approached, the

precipitation in the southern region of Kinki

district was rapidly increased. From 14:00 to 15:00, the distributions of the precipitation

(d)16:00

Fig.7 Distribution of temperature difference from

daily mean temperature Q

recorded about one hour after the daily minimum

121
122
became like a hook as the rainfall area surrounded

the center of the typhoon from the western side of

the typhoon. On the other hand, in the east side of

the typhoon, the precipitation was less than several

mm and there was the area where it was hardly

raining. After that, the rainfall area moved to the

northeastern direction with the movement of the

typhoon. The distribution shape of a hook became

illegible and the precipitation became 0 mm in Kinki district except the northern region.

The distributions of the specific humidity from

13:00 to 16:00 are shown in Figs.9(a)-(d). In the

figure shows the position of the atmospheric

pressure center of the typhoon at the time that was

found from the objective analysis (Fujii et al., 1999). As typhoon Vicki hit Kinki district, the

dried air entered into Kii peninsula from the

southwestern direction and caught the damped air of the northeastern side and became the

distribution like a swirl figure around the eye of

the typhoon from 14:00 to 15:00. That the specific

humidity exceeded 20g/kg in the border in Wakayama and Osaka prefecture at 14:00 seems to be brought by the rain band which was developed on the western side of the typhoon. On the other

hand, the specific humidity in the east side of the

typhoon became 10-15g/kg with the dried air flowing in from the southwestern direction of

Kinki district.

3.6 Situation of Atmospheric Pressure

Figs.lO(a)-(d) are atmospheric pressiu-e

distributions from 13:00 to 16:00. The

atmospheric pressures observed at the

meteorological observatories were the absolute

pressures revised at the surface of the sea, but the

atmospheric pressures observed at almost fire stations were not revised at the surface of the sea, and also the zero of the barometers in some fire stations was sometimes shifted because they were

not given official approval of Meteorological

Agency. Therefore, it expressed as the

(a) 13:00

(b)14:00

(c)15:00

(d)16:00

Fig. 10 Distribution of differences from maximum

atmospheric pressure before daily minimum

atmospheric pressure (hPa)

123
atmospheric pressure difference between the

atmospheric pressure at each time and the

maximum atmospheric pressure before the passage

of the typhoon. In the figure shows the position of

the atmospheric pressure center of the typhoon at the time that was found from the objective analysis

(Fujii et al., 1999). The shape of the distribution of

the atmospheric pressure was almost a circle at 14:00 and an elliptical form at 13:00, 15:00

andl6:00.

4. RELATION BETWEEN COMBINED

RADAR ECHO AND SEVERE STORM IN

KINKI DISTRICT

Figs.l l(a)-(p) are the combined radar echo charts

by every 7 minutes and 30 seconds from 14:00 to 16:00 and the daily maximum instantaneous wind

velocities were displayed as vectors in these

figures according to the time when the daily

maximum instantaneous wind velocity was

recorded in each place. The combined radar echo

of 14:45 lacks data. Described in Fig.6, the severe storm area is mainly divided into two areas where

the minimum atmosphere pressures and the daily

maximum instantaneous wind velocities were

recorded at the almost same time and the daily

maximum instantaneous wind velocities were

recorded about one hour after the minimum

atmosphere pressures occurred. The former

corresponds to the severe storm area that occurs in the right side of the typhoon whereas the latter is the severe storm area formed behind the center of

the typhoon. Yamamoto et al.(1963) reported the severe storms accompanied with the cold front in their report of 2nd Muroto typhoon. This severe storm behind typhoon Vicki is supposed to have

the same character as the severe storms observed

in 2nd Muroto typhoon. There are some reports of

observation (Nakajima et al., 1980 and Fujii et al., 1992) on this severe storm but it seems that there

is not detailed observation report yet.

In the combined radar echo chart at 14:00, there is a rain band with strong rainfall on the western side

of the typhoon. The severe storms occurred in the

direction of the rotation of the typhoon inside the

rain band at the both ends of north and south at the

rain band. The severe storm in the west-northwest was observed inside the rain band in Sumoto city,

Hyogo prefecture at 14:15. The south severe storm was observed by Matsuzaka city, Mie prefecture at the same time but this severe storm occurred in the

right side of the typhoon. The west severe storm was also observed in the Wakayama city. This

severe storm was observed 6 minutes later after at Sumoto city. These severe storms were observed at Sumoto and Wakayama city inside the same rain

band. After that, the severe storm area moved to the northwestern west with the movement of this rain band. At 14:37 the severe storm area moved

from the northern region of Wakayama prefecture

to the western region of Nara prefecture.

The two severe storms were observed in the south

direction in the Nara basin and in west-southwest

direction along Yoshino River. The wind at Yoshino River had the direction of the rotation of

the typhoon whereas the south wind on the Nara

basin became the direction of the normal of the

typhoon. The south severe storms occurred to 15:00 across Mt. Kongo located in the border

between Osaka and Nara prefecture. The

maximum wind velocity was observed 59.5m/s at Shinjo town in Nara prefecture. Only the wind

direction at Shinjo town was south-southeast.

After that, at 15:07, the almost west severe storms were observed around this rain band from the

southem region of Osaka prefecture to the

northern region of Nara prefecture. In this way

within only 15 minutes from 14:52 to 15:07, the

wind direction changed from west to south and

from south to west again across Mt. Kongo. The

author's field study in Nara prefecture got a lot of

testimonies, that at first the south wind blew and

124
125
(p)I6K)0

(continued)

126
then changed to the west wind, and a lot of

evidences of the damages with the west wind in addition to the south wind. It thinks that the severe storm with such the rapid change of the wind

direction occurred. This phenomenon seems to be

caused by the meteorological factors as well as the

geographical features. Comparing with the

temperature distributions at 14:00 and 15:00 and

these combined radar echo charts, it finds that this rain band was approximately on the 0 degree

isotherm on the ground. That is, the severe storm

occurred in the neighborhood of the 0 degree

isotherm. It thinks that the temperature

distributions related to the occurrence of the severe storms.

Sakakibara et al.(1999) showed that the

converging shear line existed near the rain band

over Nara city fi^om the results of the radar echo

observed at Kansai international airport at 15:30.

They reported that this converging shear line caused the occurrence of severe storm on the

ground. Kawano et al.(1999) observed 3

components of the wind velocity in the typhoon

Vicki using MU radar system in Shigaraki town,

Shiga prefecture. At 15:53, their observation

stopped by the blackout but they observed the

strong down stream, 4-5m/s, in the space of 2-3

km high over Shigaraki town. At the same time,

the rain band above-mentioned approached over

Shigaraki town fi^om the results of the combined

radar echo charts. The daily maximum

instantaneous wind velocity was observed 36.4m/s

at 15:45, in Ritto town and 41.6m/s at 16:16 in Yoka-ichi city in Shiga prefecture. That is, it thinks that there is some causality between this down stream and the severe storm on the ground.

Using the weather observation data at the fire stations, the interesting facts were found about the severe storm behind typhoon Vicki. The weather

elements in the sky of the typhoon can be

observed quite detailed using the radar echo

systems or the meteorological satellite named

Himawari. But the weather elements on the

ground have been observed only by

meteorological observatories or AMeDAS and

there have been few amounts of weather

information on the ground. However, using the

weather observation network at the fire stations in addition to the meteorological observatories and

AMeDAS, it is possible to expect that the more

detailed weather observation on the ground of the

typhoon becomes possible.

5. CONCLUSION

Mainly using the weather observation data at fire stations, it investigated about ground weather

elements in typhoon Vicki. The results as follow:

1) It foimd that the weather observation network

at the fire stations were extremely useful

because the weather observation network at the fire stations were much closer than the

observation network of the meteorological

observatories and AMeDAS in the urban

region.

2) The three courses of typhoon Vicki on Kinki

district was showed meandering. Especially,

the wind center and the atmospheric pressure

center resembled very well each other. 3) The severe storm area in typhoon Vicki was

divided into two areas, which were formed in the right of the typhoon and in the rear of the

typhoon. The severe storm of the former was

observed in Wakayama prefecture and Aichi

prefecture. The severe storm of the latter was

observed in the central region of Kinki district and brought enormous damages to each place

in Kinki district. 4) The distribution of the temperature showed

that the cold air entered into the center of the

typhoon from the northwestern side. It was

foimd that the temperature gradient in space

became very large near the prefecture border

between Osaka and Nara prefecture at 15:00.

5) The distribution of the hourly precipitation

showed the rainfall area in typhoon Vicki.

6) The distribution of specific humidity showed

that the dried air from the southwestem

direction and damped air from the

northeastern direction the state flew into the center of the typhoon at 15:00.

7) The shape of typhoon Vicki was often an

elliptical form shown by the distribution of

the atmospheric pressure. REFERENCES

127
Fujii T. et al.(1992): On Analysis of Typhoon

9119 and Distribution of Severe, Annuals of

Disaster Prevention Research Institute, Kyoto

University, No. 35 B-1, pp.l83-191(in

Japanese).

Fujii T. et al. (1999): Relation between Wind

Velocity Calculated from Atmospheric Pressure

Field and Observed Wind Velocity (Part 3)-

Analysis of Typhoon Vicki, Annuals of Disaster

Prevention Research Institute, Kyoto

University, No. 42 B-1 (in Japanese).

Kawano N. et al.(1999): Observation of Typhoon

Vicki Used a Method of MU radar, Preprint of

annual scientific meetings, Meteorological

Society of Japan, pp.223 (in Japanese).

Meteorological Agency (1998a): Ground Weather

Observation Data, Area Weather Observation

Data and Upper-Air Observation Data,

Meteorological Agency Monthly ReportSeptember, (in Japanese).

Meteorological Agency (1998b): Heavy Rain and

Severe storm by Typhoon No.6, No. 7, No. 8,

No. 9 and front from 18th September to 2nd

October, Disaster Weather Newsletter

Nakajima C. et al.(1980): On Typhoon 7916,

Annuals of Disaster Prevention Research

Institute, Kyoto University, No. 23 B-1, pp.87-

111 (in Japanese).

Sakakibara H. et al(1999): Severe Storm Caused

by Typhoon Vicki, Preprint of annual scientific meetings. Meteorological Society of Japan,

pp.215(in Japanese).

Shaw N.(1918): The Travel of Circular

Depressions and Tornadoes, Meteorological

Office, Geophysical Memoirs, No. 12, pp. 19-44.

Uematsu Y. et al.(1999): Wind Measurements in the Tohoku District (Part 1 Aomori and Akita

prefectures). Journal of Wind Engineering, No.

78, pp.8 1-93 (in Japanese).

Yamamoto R. et al.(1963): On Distribution of

Severe Storm of 2nd Muroto Typhoon, Annuals

of Disaster Prevention Research Institute,

Kyoto University, No. 6, pp.ll3-I27(in

Japanese).

128
Theme 3

Earthquake

Engineering

129

Seismic Testing of a 1/20-Scale 2D Model of Koyna Dam

by

Robert L. Hall', Mostafiz R. Chowdhury', and Enrique E. Matheir

ABSTRACT

A series of shaking table experiments were

proposed as part of the ongoing earthquake

engineering research program at the U.S. Army

Engineer Research and Development Center

(ERDC) as an effective way to provide realistic data for validation of numerical procedures

modeling the seismic performance of concrete

gravity dams. This paper describes the first experiment of the series, performed on a 1/20-

scale model of Koyna Dam. The objective of

this experiment was to study the dynamic

response characteristics of the model when

subjected to a sinusoidal horizontal base

excitation acting on the upstream-downstream

direction (in-plane or transversal direction). The

value of the excitation frequency was selected

slightly higher than the first in-plane natural

frequency of the model. The amplitude of the

excitation was increased for each test run until failure occurred at an excitation level of 0T6g.

The model behavior was documented with

acceleration, displacement, and strain

measurements. The recorded data will provide

valuable information for the update and

calibration of linear and nonlinear numerical

models predicting the seismic response of

concrete gravity dams.

KEYWORDS: Concrete dams, scaled model

experiments, nonlinear response.

1. INTRODUCTION

The seismic research and development in the area of concrete dams is currently focused on the

generation and validation of three-dimensional

time-domain analytical procedures to predict the

dynamic performance of these critical structures. The main objective of the procedures under

development is to capture the most important

phenomenological aspects controlling the

dynamic performance of the structural system,

including detailed representations of the

interactions between the subsystems (dam,

foundation, reservoir, and sediments) and the

nonlinear characteristics of the material

behavior. As part of the ongoing earthquake

engineering research program at the U.S. Army

Engineer Research and Development Center

(ERDC), a series of shaking table experiments

have been formulated as an effective way to provide realistic data for validation of numerical

procedures modeling the seismic performance of

concrete gravity dams.

This paper describes the first experiment of the

series, performed on a 1/20-scale model of the

tallest nonoverflow monolith of Koyna Dam.

This 103m high concrete gravity dam is one of

the few concrete dams to have experienced

significant damage induced by seismic ground

motions. The dam is located on the Koyna River

in the western 'side of India, and it was

completed in 1963. The structure is comprised

of 56 monoliths, about 15.24m wide each (Saini

et al., 1972). The 1967 Koyna earthquake, with

magnitude of 6.5 and epicenter in the immediate

vicinity of the dam, caused significant structural damage, including horizontal cracks on the

upstream and downstream faces of a number of

nonoverflow monoliths around the elevation at which the slope of the downstream face changes

abruptly. The overflow monoliths were not

damaged. The earthquake was recorded on a

strong motion accelerograph located in a gallery

of the dam. Although the structure did not appear to be in danger of imminent catastrophic

failure, the damage induced by the earthquake was serious enough to require permanent

strengthening by adding concrete buttresses on

the downstream face of the nonoverflow

monoliths (Chopra, 1978).

' Structures Laboratory, U.S. Army Engineer Research and Development Center, Vicksburg, MS 39180 (USA).

^ Department of Civil and Environmental Engineering, Louisiana State University, Baton Rouge, LA 70803 (USA).

131
The historic performance of Koyna Dam and the

special characteristics of its cross section have

resulted in this structure being a classical

problem for experimental studies and for the

validation of numerical procedures for predicting the seismic response of concrete

gravity dams. Previous tests performed by the

U.S. Army Corps of Engineers, the University of

California at Berkeley, and the Bureau of

Reclamation can be cited among the

experimental studies performed in the U.S.

which were based on scale models of Koyna

Dam. These tests are discussed briefly in the

following.

A comprehensive series of experimental tests involving a scale model of this gravity dam were

performed at the University of California at Berkeley (Niwa and Clough, 1980). An

extensive testing program was carried out to obtain an appropriate material mix that would

allow the reduced-scale investigation of the

nonlinear response and failure characteristics of

concrete dams. A mix based on casting plaster,

celite, sand, lead power and water was adopted

for the model representing a monolith of Koyna

Dam at a 1/150 scale. The lead power was added

to approximate the density of the prototype

material. The strength characteristics of the mix

were adequate but the resulting elastic modulus was significantly higher than the assumed

prototype value. The presence of the reservoir was incorporated in the model by using a rectangular tank filled with water. The first natural frequency of the model under empty

conditions was determined to be about 43 Hz

(3.51 Hz, prototype scale). This value is higher

than the values usually obtained in numerical

studies because of the increased elastic modulus.

The original idea was to test the model using a

scaled version of the Koyna record. However, it was determined that the dynamic characteristics

of the shaking table would result in attenuation

of motion components for frequencies above 16

Hz. Therefore, because of these equipment

limitations, an artificial harmonic excitation was

used in the dynamic tests. The amplitude of the

peak table acceleration varied from 0.075g to 1.21g. During the 1.21g-level test, a crack

formed at the elevation corresponding to the

downstream slope change and it propagated

completely through the section. After cracking,

the upper part of the model subsequently

engaged in a rocking motion characterized by

large crest displacements.

Another series of dynamic tests were performed on a 1/60-scale model of Koyna Dam at the

facilities of the U. S. Army Engineer Waterways

Experiment Station (Norman, 1986). The

material properties were not scaled, and the

main objective of these tests was to investigate

the failure characteristics of a geometric model

constructed of standard low strength concrete. The presence of the reservoir was not considered

in these experiments. Forced vibration tests were

conducted to determine natural frequencies and

mode shapes. The first two natural frequencies were identified at about 3.43 Hz and 7.55 Hz

(prototype scale). The dynamic failure tests were

performed by inducing the motion of the

foundation block through the impact of a moving mass. The foundation block was placed on top of a smooth sliding surface, and it was

connected to a reaction wall through a multispring system. During the base excitation test, a crack immediately formed at the elevation

corresponding to the downstream slope change.

Unfortunately, many of the sensors went off- range during the test, and most of the

conclusions were of a qualitative nature.

The Bureau of Reclamation recently performed

a series of shaking table experiments on a 1/50-

scale model of Koyna Dam (Harris at al., 1999).

The presence of the reservoir was not included

in the modeling. A mix based on cement,

bentonite, sand, gravel and water was designed

to satisfy the similitude requirements. An

advantage of this mix is that it did not rely on

lead products to achieve the similitude

requirements regarding unit weight. Two models were built and tested. The first model was cast horizontally and lifted onto the table after 28

days. This model developed a shrinkage crack

along a cross section at about the middle of the

height. The second model was cast upright and

tested at an earlier age to avoid the cracking

experienced in the first model. Both models were tested using a 14 Hz (1.98 Hz at prototype

132
scale) sinusoidal excitation whose amplitude was gradually increased until failure. This

excitation frequency was well below the first natural frequency associated with the in-plane or

transversal response of the models, which was

about 24 Hz (3.39 Hz, prototype scale). The test runs performed on the first model exhibited the

expected nonlinear behavior caused by the

initially cracked condition. The tests on the

second model showed the abrupt development of

a crack pattern that eventually caused the

separation of the top section of the model. It was

observed that these results were influenced by

the failure of the foundation of the model, which

created a boundary condition difficult to represent in numerical models.

This paper describes a new series of shaking

table experiments performed on a 1/20-scale

model of Koyna Dam. A diverse team of

engineers and scientist from different ERDC

Laboratories collaborated in the completion of

these experiments. The special material mix

used for the model was developed at the ERDC

Structures Laboratory (Vicksburg, Mississippi),

which had the project lead responsibilities. The

tests were carried out using the Triaxial

Earthquake and Shock Simulator (TESS) at the

ERDC Construction Engineering and Research

Laboratory (Champaign, Illinois). A series of

tests were conducted to record the natural

dynamic characteristics as well as the controlled

forced response of the instrumented Koyna dam

model when subjected to a sinusoidal horizontal

base excitation acting on the upstreamdownstream direction (in-plane or transversal

direction). The frequency of the sinusoidal

excitation was selected slightly higher (14 Hz)

than the first in-plane natural frequency of the

model. Peak support accelerations were varied

from 0.005 to 0.16 g, level at which failure

occurred. This type of sinusoidal input was used

to ensure that the model response would be

dominated by motions associated with the first mode of vibration in the in-plane direction. The

recorded data included the input and output

motions including acceleration, strain, and linear

variable differential transducer (LVDT) timehistories.

2. MODEL DESCRIPTION

The purpose of small-scale modeling is to study

the response of the structural system (prototype)

by reproducing its behavior using a reduced size

version of the structure that satisfies the

constraints imposed by the equipment available

or the laboratory conditions. In order for the

model to reproduce meaningfully the behavior

of the prototype, it is necessary the satisfaction

of a set of similitude requirements relating the

physical and time scales of the model along with

the corresponding material parameters. The

number of these similitude constraints depends

on the nature and characteristics of the problem.

Let us define a generic scaling factor as follows:

^ model where a represents any variable of the problem.

In general, if a number of "basic" scaling factors

is predetermined, then the remaining scaling

factors can be determined by dimensional

analysis. The number of basic scaling factors

corresponds to number of similitude

requirements, that is, the minimum number of

independent constraints that must be satisfied to guarantee the proper representation of the

problem at the model scale. For this case, the

scaling factors corresponding to geometry

(length), modulus of elasticity and density are adopted as the basic scaling factors. They are denoted by X^, X^, and , respectively. The

scaling factors corresponding to the remaining

physical variables of the problem can be written

in terms of these factors as indicated in Table 1. The table shows the particular case in which the

density of the model material is kept the same as the density of the prototype material while the

geometry and strength variables are subjected to the same scaling. For this case, all the scaling

factors can be written in terms of the scaling

factor defining the geometry. This was the

scaling strategy adopted for this experiment,

based on a 1/20-scale model (X,^ - 20 ).

133
Table 1. Scaling relationships

Variable Scaling factors

Time

^1 =

Frequency

A

_

1

Force -

Stress

Strain K = ^

Displacement

Velocity

Acceleration

The corresponding model at the 1/20 scale is shown in Figure 1, with a height of 5.15m. It can

be noticed the abrupt change of slope on the

downstream face, at an elevation of 3.32m with

respect to the base level. The maximum

dimension at the base of the model is 3.51m.

According to the selected geometric scale, the

model width should be 0.76m. However, this could have resulted in a very weak model with

respect to the out-of-plane direction. Although

the test was designed to investigate only the in- plane or transversal behavior of the monolith,

small variations in geometry, material properties or actual test conditions could induce the

undesirable excitation of the out-of-plane

response of such a slender structure. Since the

model was deprived of the beneficial restraining

effect of the adjoining monoliths present in the

real structure, there was some concern about the

possibility of failure occurring in this unnatural

mode. It was decided therefore to investigate the

feasibility of wider models, stronger in the out- of-plane direction. Finite element analyses were

performed on three-dimensional models of the

monolith at a 1/20 scale. The objective of these

analyses was to determine the influence of the

width on the free vibration characteristics of the

model. The material properties used were as follows: unit weight, 23.57 kN/m3; modulus of

elasticity, 1.379 GPa; Poisson's ratio, 0.20.

Figure 2 shows the variation of the first two

natural frequencies as a function of the width of

the model. The fundamental frequency

associated with the out-of-plane response

increases with the width, starting at 5.52 Hz for the 0.76-m wide model. The first namral

frequency associated with in-plane response

(transverse bending) is practically not affected

by changes in the out-of-plane geometry, and it has a constant value of about 13.68 Hz. It is clear that the slender model shows the advantage

of two well-separated natural frequencies, taking

into account that the model would be tested

using a sinusoidal excitation with a frequency

close to that of the fundamental in-plane mode.

However, there were some concerns about the

inherent weakness of this thin structure. Note

that if the width is increased to produce a

stronger section, the separation between these

two natural frequencies decrease. A compromise

solution was adopted for the model width,

whose value was selected to be two times the

original width, i. e., 1.52 m.

3. MIX DESIGN

A material mix was developed at the ERDC

Structures Laboratory to satisfy the similitude

requirements. The resulting material should

exhibit a density equal to that of the prototype

material, while the strength and modulus should

be 1/20 of the corresponding prototype values.

Note that the density requirement translates into

unity scaling for the unit weight. Table 2 lists the material properties assumed for the

prototype material, and the corresponding target

values for the model material.

Table 2. Material properties

Property Prototype Model

Unit weight 23.57 kN/m^ 23.57 kN/m'

Modulus of elasticity 27.58 GPa 1.379 GPa

Compressive strength 27.58 MPa 1.379 MPa

Tensile strength 2.76 MPa 0.138 MPa

Poisson's ratio 0.20 0.20

Several trail mixes were tested in the laboratory

in order to design a material mix that exhibited

134
reduced strength and elastic modulus while

keeping the values of density and ultimate

compressive strain close to those considered

standard for normal concrete. The resulting mix was a very fluid paste, characterized by a water- cement ratio of 2.95, which was intended to produce a low strength material. The main

aggregate of the mix was barite powder, which

is a material with a high specific gravity and it was added to control the value of the unit

weight. A small amount of plastic fibers was

also incorporated in the mix to improve the

ductility and deformation characteristics of the

material. Table 3 shows the composition of the

final laboratory mix.

Table 3. Mix composition

Material Specific

Gravity

Volume

(1 m')

Cement 3.15 0.0536

Natural concrete sand 2.56 0.0833

Barite 4.25 0.3501

Plastic fibers 0.91 0.0050

Water 1.00 0.4980

The mix was commercially ordered and

supplied, and the material was batched in readymix trucks. Standard control procedures were

followed to ensure the quality of the material,

and laboratory testing was performed on test cylinders and beams. The average values of the

material properties corresponding to the

specimens tested at an age of 28 days are shovm

in Table 4. The resulting material was stronger

than the original laboratory mix, and it exhibited

a compressive strength 1.7 times the original

target value. The flexural modulus of elasticity,

obtained from third-point loading testing on beam specimens, was about 2.4 times the target

value. Note also that the ratio between the

splitting tensile strength and the compressive

strength is 0.12, which is close to the normal

value for the same ratio in standard low-strength

concrete.

After casting, the material showed significant

settlement, and the amount of bleeding was

higher than anticipated from the laboratory

material tests.' The significant bleeding caused

the model to be slightly shorter than the original

design.

Table 4. 28-day material test results

Property 28-day value

Unit weight 24.33 kN/m^

Modulus of elasticity (flexural) 0.329 GPa

Compressive strength 2.38 MPa

Tensile strength 0.29 MPa

Modulus of rupture 0.69 MPa

Several drilled core samples were taken from the

model after the completion of the dynamic tests. The cylindrical core specimens were obtained

from the material near the critical section

(elevation corresponding to the downstream

slope change) and near the base of the model.

The samples were taken from these two different

locations in order to quantify the variation of

material properties along the height of the

model. The core specimens were tested at an age

of 1 1 5 days, and the resuhs of the corresponding

tests are shown in Table 5.

Table 5. 115-day material core tests

Property a) Base b) Critical section

Unit weight 23.28 kN/m^ 19.56 kN/m'

Modulus of elasticity 3.447 GPa 1.379 GPa

Compressive strength 4.60 MPa 1.07 MPa

Tensile strength 0.60 MPa a/a

It is clear that the properties of the material were

not homogeneously distributed along the height.

The material near the base showed strength and

elastic modulus values that were significantly

higher than those predicted from the previous

laboratory tests. On the other hand, the material near the critical location was characterized by

reduced density and strength values. Visual

inspection also confirmed the inhomogeneous

characteristics of the material. In particular, the

fiber content seemed to be abnormally low in those specimens taken from the region near the

critical section. A possible explanation for these

135
results could be based on the exaggerated

settlement of heavier solid particles that trapped

the fibers forcing them to migrate towards the

base of the model. This phenomenon could have

been fueled by the fact that the barite aggregate

used for the model was obtained from a different

provider than the one used for the laboratory

mix in the design phase.

The dam model was anchored to the shaking

table by means of a reinforced concrete base

beam with a height of 0.46m and with

reinforcing steel that extended 0.61m inside the

model. Instrumentation included accelerometers,

strain gauges, and linear variable differential

transducers (LVDTs) installed on the model

surface for a total of 56 data channels.

Accelerometers were attached to the base beam

to characterize the input motions at the base of

the model. Accelerometers were also installed

along the height of the upstream face in a 3- column pattern, to check for any undesired

torsional response. Strain gauges and LVDTs

were installed near the expected location for crack initiation, at the change of slope along the

downstream face.

4. SHAKING TABLE TESTS

Modal tests : Prior to conducting any tests using

the shaking table to excite the model, modal

tests were conducted using an electrostatic

shaker attached to the upstream face of the dam

model. The model response was recorded using a scaiming laser vibrometer system. The

vibrometer was located directly in front of the

upstream face, with the laser head positioned at the average elevation of the model upstream

surface. The response to the random excitation

provided by the shaker was recorded by

scanning the upstream face of the model

according to a predetermined grid pattern. From

these tests, the first two natural frequencies in the in-plane direction were measured at 13.4 and

34.4 Hz, respectively. The values of the

corresponding damping ratios were identified as

5.5 and 2.5%, respectively.

Sine-sweep tests : A series of sine-sweep tests were performed using the shaking table to

determine the in-plane vibration characteristics

of the model. These tests began a frequency of 4

Hz and swept up to 64 Hz (4 octaves) at a rate of

5 seconds per octave, with the exception of the

third test of the series that was conducted at a slower rate of 10 seconds per octave. The tests began at a low excitation level, which was

gradually increased until the model was

sufficiently excited to record modal parameters

with good data resolution. The test at the slower

sweep rate provided the cleanest signals and it was used to estimate the model natural

frequencies. A transfer function was computed

between the horizontal components of table

acceleration and model crest acceleration. Figure

3 shows the resulting transfer function. The

natural frequencies corresponding to the first and second in-plane vibration modes can be

estimated from the magnitude peaks as 12.7 and

37.0 Hz, respectively.

Koyna time-history test : Before the sinusoidal

tests, the Koyna model was tested using as input

a scaled version of the transversal component of

the 1967 earthquake. The peak acceleration

corresponding to the original record was about

0.3 8g. To avoid inducing any damage to the

model, the amplitude of the acceleration record was scaled down to 0.027g. Figure 4 shows the

resulting input motion. The variation of the peak

values of absolute acceleration response along

the upstream face is summarized in Table 6, in which the elevation of each sensor is defined in terms of the total height H of the model. The

accelerometer A9x, located along the upstreamface centerline, did not work properly, and

therefore the acceleration at the corresponding

elevation is reported in terms of A14x, located at the same elevation but close to a lateral face. Table 6. Peak absolute acceleration values

Sensor Elevation Koyna SineS

A7x 1.5% H 0.027g 0.136g

A8x 25.7% H 0.049g 0.214g

A14x 65.4% H 0.073g 0.568g

AlOx 98.5% H 0.204g 1.131g

136
Sinusoidal tests : The model was tested to failure

with sinusoidal motions with a frequency of 14

Hz. This value was selected slightly above the

fundamental in-plane natural frequency. This was done to avoid the possibility of a complete

resonant condition arising from a shift of the

fundamental frequency induced by damage.

Each of the sinusoidal support motions ramped

up to the maximum amplitude level in 1 second,

maintained this level for 5 seconds, and then

ramped down to zero in 1 second, for a total duration of 7 seconds. The series of sinusoidal

tests began at very low amplitude levels (0.005 g

for the first test in the series). Test levels were

gradually increased, and the response of the

model remained essentially linear up through the

fifth test in the series (SineS), with maximum

amplitude of 0.1 2g. Figure 5 shows the time

history of the base-beam horizontal acceleration

corresponding to this test. Figure 6 shows the

time history of the acceleration at the top of the

model, measured on the upstream face. Figure 7

shows a particular strain time history near the

critical location, recorded by a strain gauge

located 2.5 cm above the change of slope along

the downstream face. The peak acceleration

profile is summarized in the last column of

Table 6.

The last test in the series (Sine6), with maximum

amplitude of 0.1 6g, induced the first nonlinear

response and led to the ultimate failure of the

model. Figure 8 shows the time history of the

crest acceleration. Figure 9 depicts the strain history at the critical downstream location,

indicating the initiation of tensile cracking at approximately 1.27 seconds. According to the

other measurements near the critical location,

the crack affected completely the downstream

face during the next two response cycles (at 1.41 seconds). This primary crack propagated to the

upstream face at 1.52 seconds, at an inclined

angle, consistent with previous experimental

observations. Figure 10 shows detailed views of

the damaged area.

5. CONCLUSIONS

A systematic validation and updating of an

analytical finite-element model using benchmark

experimental' data is essential for the

development of representative, reliable, and

credible virtual tools predicting the response of

physical structures. In spite of considerable

research efforts towards modeling the dynamic

response of large concrete dams, only a limited

number of well-documented correlation studies

using experimental data are available.

This study was based on a large 1/20-scale

model that was developed to minimize the

modeling distortions usually associated with

small-scale models. A new concrete mix

consisting of cement, plastic fibers, barite, sand,

and water was designed based on the governing

scaling relationships. The dynamic behavior of

the model was extensively documented. The

model failed exactly as predicted by the field observations, with a crack beginning at the

change in slope on its downstream face. The

recorded data will provide valuable information

for the calibration of numerical models

predicting the seismic response of concrete

gravity dams.

6. ACKNOWLEDGEMENTS

This work was sponsored by the Direct-Allotted

Civil Works Earthquake Engineering Research

Program of the U.S. Army Engineer Research

and Development Center in Vicksburg,

Mississippi. The authors appreciate the

cooperation of the authorities at the U.S. Army

Engineer Research and Development Center,

which permitted the presentation and

presentation of this study. Permission to publish

this paper was granted by the Chief of

Engineers, U.S. Army Corps of Engineers.

7. REFERENCES

Saini, S. S., Krishna, J., and Chandrasekaran, A.

R. (1972). "Behavior of Koyna Dam - Dec. 11,

1967 Earthquake," Journal of the Structural

Division, ASCE, Vol. 98, No. 7.

Chopra, A. K. and Chakrabarti, P. (1971). "The

Koyna Earthquake of December 11, 1967 and

the Performance of Koyna Dam," Report No.

EERC 71-1, Earthquake Engineering Research

137
Center, University of California at Berkeley,

CA.

Chopra, A. K. and Chakrabarti, P. (1973). "The

Koyna Earthquake and the Damage to Koyna

Dam," Bulletin of the Seismological Society of

America, Vol. 63, No. 2, pp. 381-397.

Chopra, A. K. (1978). "Earthquake Resistant

Design of Concrete Gravity Dams," Journal of

the Structural Division, ASCE, Vol. 104, No.

ST6, pp. 953-971.

Hall, J. F. (1988). "The Dynamic and

Earthquake Behavior of Concrete Dams: Review

of Experimental Behavior and Observational

Evidence," Soil Dynamics and Earthquake

Engineering, Vol. 7,' No. 2, pp. 58-121.

Norman, C. D. (1986). "Dynamic Failure Tests and Analysis of a Model Concrete Dam,"

Technical Report SL-86-33, Structures Laboratory, USAE Waterways Experiment

Station, Vicksburg, MS.

Harris, D. W., Snorteland, N., Dolen T., and

Travers, F. (1999). "Shaking Table 2-D Models

of a Concrete Gravity Dam for Computer Code

Validation," Report DSO-98-13, Bureau of

Reclamation, Dam Safety Office, Denver, CO.

138
1.52m

3.51mFigure 1 . Model

0.00 0.25 0 76 1 00 1.25 l.SO Time [sec]

Figure 5. Base beam acceleration (SineS)

0.50 1.25 1.50 1.75 2.00 Monolith width [m]

Figure 2. Variation of model frequencies

0.75 1.00 1.25 Time [sec]

Figure 6. Crest acceleration (Sine5)

0 03

002

0.01 c

_o

rat 0.00 o

o

u -001 o<

-0.02

-003

30 40 50 Frequency [Hz]

Figure 3. Transfer function

270

125 210 100 150 75

90 05

0) 30 S

o

-30 <n

CO

"Sc

ro

5025

0

-25 -90 a 35 -SO -150 -75

-210 -100

-270 -125

1

000 0 25 0 50 0.75 1.00 1.25 1.50 1.75 2,00 2,25 2 50 Time [sec] Figure 4. Koyna scaled record

0.75 1.00 1 25 Time [sec]

Figure 7. Critical strain (SineS)

0 75 1.00 1 25 Time [sec] Figure 8. Crest acceleration (Sine6)

139
Figure 9. Critical strain (Sine6)

Figure 10. Damage after failure (Sine6)

140
Development of Smart Systems for Building Structures

by

Shunsuke OTANI \ Hisahiro HIRAISHI \ Mitumasa MIDORIKAWA \

Masaomi TESHIGAWARA ^ Hideo FUJITANI ^ Taiki SAITO "

ABSTRACT

Building Research Institute, Japanese Ministry

of Construction, initiated a 5-year research and

development project of "Smart Materials and

Structural Systems" in 1998 as a part of U.S.-

Japan cooperative research efforts. The U.S.

Counterpart is the National Science Foundation.

Smart Structural Systems (also called as Autoadaptive Media) are defined as systems that can automatically adjust structural

characteristics, in response to the change in external disturbance and environments, toward

structural safety and serviceability as well as the

extension of structural service life. The research

and development of (1) concept and

performance evaluation of smart structure system, (2) sensing of structure performance,

and (3) development and evaluation of structural elements using smart materials will be

conducted.

Keywords: Smart Materials, Smart Structural

Systems, Autoadaptive Media, Sensing

Technology

1. INTRODUCTION

A conventional structural system is designed to achieve a set of intended functions under preselected loads and forces. Such a conventional

system can not successfully develop its ability

against unexpected loads and forces imless a

large safety factor is provided for safety limit

states to take into account various uncertainties

in load and force amplitudes and structural response. Furthermore, since seismic design

requirements have been improved after each

lessons learned through past earthquake

disasters, the safety level of old buildings are always inferior to new buildings as evidenced in many past earthquake disasters, e.g., the 1995

the Great Hanshin-Awaji earthquake disaster.

Strengthening or removal of those old buildings

becomes necessary to protect societal welfare.

Smart Structural Systems are defined as

structural systems with a certain-level of

autonomy relying on the embedded functions of

sensors, actuators and processors, that can

automatically adjust structural characteristics, in response to the change in external disturbance

and environments, toward structural safety and

serviceability as well as the extension of

structural service life.

The Building Research Institute (BRI), Ministry

of Construction, Japan and the U.S. National

Science Foundation (NSF) initiated the U.S.- Japan Cooperative Research Program on

Autoadaptive Media (Smart Structural Systems)

in 1998, under the aegis of the U.S.-Japan Panel on Wind and Seismic Effects of the U.S.-Japan

Cooperative Program in Natural Resources. The

First Joint Technical Coordinating Committee

(JTCC) Meeting was held in Tsukuba, Japan,

from January 6 through 8, 2000. At the meeting,

research items and plans are discussed in detail corresponding to three research thrusts: (1) structural systems, (2) sensing and monitoring

technology, and (3) effector technology.

This paper described the research items and

plans of the research program in Japan based on

the recommendations summarized at the JTCC

meeting. To achieve research objectives,

following three sub-committees have been

formed under Technical Coordinating

^ Department of Architecture, University of

Tokyo, Tokyo, Japan

*' Building Research Institute, Ministry of

Construction, Tsukuba 305-0802, Japan

141
Committee of the project, chaired by Prof. S. Otani, University of Tokyo:

"Sub-committee on structural systems" chaired

by Prof. A. Wada, Tokyo Institute of

Technology,

"Sub-committee on sensing and monitoring

technology" chaired by Prof Y. Kitagawa,

Hiroshima University,

"Sub-committee on effector technology" chaired

by Prof T. Fujita, Institute of Industrial Science,

University of Tokyo.

2. CONCEPT OF SMART STRUCTURAL

SYSTEMS FOR BUILDINGS

The concept of smart structural system was

initially proposed in the field of aerospace

engineering, however, structural systems for building engineering have different features

from those for aerospace engineering as shown

in Table 1 . The value of a building should be

determined not only by structural safety but also

taking into account non-engineering points of

view such as "beauty", "economy", and

"function." Therefore, in the first stage of

research project, the objectives and needs of

smart structural systems for building

engineering should be clarified. Table

2 summarizes the research needs of a smart

structural system for buildings.

Based on the conceptual study of smart

structural systems for buildings, the following

items are selected as research targets for the subcommittee on structvual systems (chaired by

Prof. A. Wada, Tokyo Institute of Technology).

(a) Autoadaptive structural systems

Proposal of feasible structural systems with

autoadaptive features

Integration of smart materials or devices / members to achieve target performance of

structural systems

Investigation of the proposed structural systems

by computer or experimental simulations

Establishment of performance evaluation

guidelines for structural systems with

autoadaptive features

(b) Reinforced Concrete (RC) structural systems

with damage fuses

Examination of performance based design/

assessment methods

Selection of damage-control devices and

damage-sensor systems

Development of performance based design/

assessment methods

Development of repairing/ replacing techniques

for damage-control devices and damage-sensor

systems

(c) Innovative life safety systems

Survey of causes of collapse of traditional light

construction

Proposal of innovative life safety systems for traditional light construction

Investigation of proposed systems* by computer

and experimental simulations

Establishment of performance guidelines for life safety systems in traditional light construction

* such as using aluminum alloy window

frames, reinforced sliding doors, the air-bag

system and support frame systems.

3. SENSING AND MONITORING

TECHNOLOGIES

3.1. Objectives

The purpose is to identify research needs for developing new sensing systems, which consist

of advanced sensors and procedures for data

collection, management, and interpretation.

Research is needed to develop new types of

sensors and obtain information about the longterm viability of the sensors. The sensor is one

part of smart building structures, which include

sensors, effectors, and processors.

142
Highly-reliable monitoring systems which

combine various advanced sensors with system

identification and damage diagnosis techniques

are required. The monitoring systems must be

able to pinpoint the location and determine the extent of the damage for different types of

structural systems.

3.2. Research Items

Items of research and development on the issue

of sensing and monitoring technologies are described below.

1) Sensor technology

New sensors should be wireless, portable, self contained, capable of discriminating among

different types of data, and subjected to environmental testing.

New sensors are needed to detect cracks, to monitor stress, strain, corrosion potential, and

temperature.

Innovative technologies, such as MEMS, should

be considered for new sensing devices.

2) Monitoring and Damage Assessment

Monitoring technologies that integrate the

output from global and local sensors are needed

to identify damage in real time.

New algorithms are needed to quantify the

extent of damage from ambient vibration

measurements and external excitation.

Software for efficient signal processing and

integration of data fi^om various sensors must be

developed.

Evaluation strategy of structural performance for

incorporation into smart structural systems is also required.

3) Networking

Integration of network systems to evaluate data

from multiple structiu-es throughout a city is required.

Because the life of the sensors is expected to be

longer than the life of current computer systems

or software, the interface between sensors and

monitoring systems must be flexible.

3.3. Current Activities

Ongoing research plan in the sub-committee on

sensing and monitoring technology (chaired by

Prof. Y. Kitagawa, Hiroshima University) are

classified into two subjects. One is defined as

structural health monitoring system. The other

is networking systems integrated with sensors. Current activities of two subjects are shown

below.

1) Experimental test for structural health

monitoring systems

The main target of health monitoring systems is to detect damage parts and evaluate present

structural performance. A five-story steel frame

with approximately one-third scale (as shown in Figure 2) was built to examine various damage

detection methods. Structural damage is simulated by removing members or changing

member properties in localized region. In addition to the conventional sensors to measure

floor displacements and accelerations, optical

fiber sensors are put on the surface of members

to detect local deformations. Using these

sensing data, several damage detection methods

such as the method using vibration mode shapes

are applied to evaluate the simulated damage

and the limitation and improving points of the

methods are examined.

2) Concept of sensor networking systems

It is recognized that integration of network

systems to evaluate damage data of structures throughout a city is essential to identify damage

regions and make necessary coimtermeasures.

The networking systems can be classified into

three scales; a system for a single building, a system to cover several buildings in a block, and

a system to cover all buildings in a city. The

sensor network system should be equipped with

database of structural information such as design

drawings. Also, the system should utilize effectively the up-to-date information

143
technologies such as wireless data transportation

using mobile phones and satellite image, virtual

reality techniques to visualize damage situation,

high speed internet for data transfer, etc.

4. EFFECTOR TECHNOLOGIES

4.1. Objectives

The purposes are to develop smart materials and

smart devices, and to verify that they can

effectively control responses of structures and

increase the safety and serviceability of

structural systems, and disaster recovery. At the

first stage, characteristics of smart materials and

devices will be investigated. By relating these

characteristics to their applications, methods of

implementing this technology will be evaluated,

e.g. suitable placement, required capacity,

effective use, etc. , so that engineers can design more damage tolerant structural systems.

4.2. Research Items

Items of research and development on the issue

of effector technologies are described below.

1) Smart materials and devices

The following smart materials and devices are mainly considered.

Shape memory alloys

Controllable fluids (e.g. electro-rheological and

magneto-rheological fluids)

Electro-strictive elements (e.g. piezo-ceramics)

and magneto-strictive elements

High-performance cementitious composites

(HPCC)

2) Assessment of the applicability and efficacy

of smart materials to structural systems

Large scale prototype and components for auto- adaptive devices and smartma terial

components should be developed and evaluated.

This process should entail large scale testing on

devices, on shaking table systems, and with

pseudo-dynamic methods.

The result of this testing should be used to determine dynamic mechanical properties of

auto adaptive devices and smart structural components, and of the overall structural performance of complete auto adaptive systems.

Corresponding to the listed four smart materials,

four different working groups were formulated

under the sub-committee on effector technology

(chaired by Prof. T. Fujita, Institute of Industrial

Science, University of Tokyo).

4.3. Current Activities

4.3.1. Shape memory alloys (SMA)

Shape Memory Alloy (SMA) shows three

different characteristics depending on the

temperature; shape memory effect, pseudo

elasticity and partial pseudo elasticity. The

objectives of this study are to utilize SMA for the smart structural members in order to realize

the smart structural system in buildings. Some

SMA devices will be developed for these

purposes, and the guidelines on SMA for structural design will also be summarized. The

current activities of the SMA working group

(chaired by Prof. Y. Kitagawa, Hiroshima

University) are described below.

1) Survey on properties and current application

to buildings

General properties of SMA compounded of

nickel and titanium were surveyed using

references; shape memory effect and pseudo

elasticity, transformation temperature, general

stress-strain relation and its strain rate were

quantitatively grasped. Especially, it was found

to be important to apply the SMA to structural members of buildings that the stress-strain relation depends on the transition temperature,

number and amplitude of cyclic loading, strain

rate, and surroimding temperature. At present, it

is very difficult to weld SMA with other metals,

144
but it is possible to use mechanical joints

because SMA can be drilled, cut and sliced. Table 3 summarized some examples of

application of SMA in Japan.

2) Test of SMA specimens

The required specification of SMA specimens was selected to be a wire and a bar with pseudo

elasticity under room temperature (around zero

to 30 degree centigrade). The casting variables

are reduction ratio at the cold forming and heat

temperature for shape-memory. Some

specimens were produced under the combination

of these variables and examined to compare

their stress-strain relation in tension. A part of

test results is shown in the Figures 3 and 4. A

wire with 30% of reduction ratio shows good

pseudo elasticity under room temperature.

Especially the treatment under 550 degree

centigrade heat and quench was found to give a

fine pseudo-elasticity to the wire specimen. Bars

with 1 6.5mm of diameter will be shaped without

cold form process and be memorized with

controlling heat and quench because of the

production lines.

3) Cutting and shaping of SMA bars

The various conditions of cutting external

threads and shaping SMA bars were examined.

The bits for normal carbon steel such as SS400

of JIS (Japan Industrial Standard) were found to be imsuitable for SMA bars. Now, the bits for

hard metals are tested to cut and shape the SMA

bars.

43.2. Controllable fluids (e.g. electrorheological and magneto-rheological fluids

(ER/MR))

Electro/Magneto Rheological (ER/MR) Fluids

have essential characteristics that change fi^om

free-flowing, linear viscous fluid, to a semisolid

with a controllable yield strength in milliseconds

when exposed to an electric and magnetic field. These fluids are effective materials for development of controllable devices. It is intended to develop a structure that changes its stiffiiess and damping characteristic to behave

adaptively against earthquake or wind forces

and achieve safety and functions by using

ER/MR devices with less energy. The research

items of the ER/MR working group (chaired by

Prof S. Soda, Waseda University) are described

below.

1) Quantification of mechanical characteristics

and material properties of ER/MR fluid

The mechanical characteristics are investigated

using MR fluids developed and the required

performance of new MR fluids is discussed.

2) Performance evaluation and development of

ER/MR devices

The mechanical characteristics of ER/MR

devices are investigated by experimental tests of

ER/MR devices which are available on the

market and the performance evaluation method

and the required performance of ER/MR devices

are discussed.

3) Development of adaptive structures using

ER/MR devices

Ideas of adaptive structures using ER/MR

devices are summarized and their effectiveness

will be verified by numerical simulations and

real-scale experimental tests. The evaluation

method and design method of adaptive

structures will be discussed. Examples of

application of ER/MR devices are presented in Figures 5 and 6.

4.33. Electro-strictive and magneto-strictive

elements

Electro/magneto-strictive elements (induced

strain actuator (ISA)) can change their own

shapes according to external electric/magnetic

fields, and vice versa. Recently these materials

have been widely used for the small/precision

machines because of some advantages from

viewpoint of small sizes, rapid reaction, high

power, high accuracy etc. The objectives in this study are to develop smart members for building

and to realize the smart, comfortable and safe

structures. Designing guidelines of ISA

materials/devices are also discussed. The

research items of the ISA working group

145
(chaired by Prof. T. Fujita, Institute of Industrial

Science, University of Tokyo) are described

below.

1) Development of smart structural members

using ISA devices

ISA materials are very suitable for structural control. We try to integrate these materials into

normal structural members such as colimins,

beams etc. to realize smart structural systems.

Example images are shown as Figures 7 and 8.

2) Using ISA as sensor materials

ISA materials can act as sensors because they cause change of electric or magnetic fields imder

deformation (see Figure 9). As schematically

shown in Figure 10, PVDF (Polyvinylidene

fluoride) sensors are suitable for membrane

structures.

3) Improvement of Acoustic Environment

Polymers based ISA films or distributed ISA

devices can control vibration mode of plane

members. Applications to music halls or dwelling partition walls are expected.

Experimental study of noise control is now

planning.

4.3.4. High-performance cementitious

composites (HPCC)

HPCC is chopped fiber reinforced

mortar/concrete micro-structurally designed

using micro-mechanical principles. HPCC

exhibits strain-hardening with superior strain

capacity, shear ductility, and extreme damage

tolerant mechanical behavior. The ultra ductile

behavior of HPCC, combined with its flexible

processing requirements, isotropic properties,

and moderate fiber volimie fiaction (typically

less than 2% depending on fiber type and

interface and matrix characteristics) make it especially suitable for critical elements in seismic applications where high performance

such as energy absorption, steel/concrete

deformation compatibility, spall resistance and

damage tolerance are required. Applications to both new structures and in retrofitting of

existing reinforced concrete structures to withstand fiiture earthquakes are considered.

HPCC elements are expected to decrease the

response and damage of building structures under external disturbances to achieve a high

level of building performance requirements.

Those requirements are not only structural safety but also reparability, serviceability, and

durability of buildings after extemal

disturbances. Stiffness, strength and ductility of

the HPCC elements can be controlled easily by

its dimensions and types of HPCC materials

used. Then it is expected to develop the suitable

energy dissipation and damage tolerant elements

for concrete structures with higher stiffness than

the other types of structures. Since the HPCC

elements will be able to dissipate the input

energy by extemal disturbances even in the

small deflection state of the buildings, response

and damage of building will be decreased.

Experiments and studies on HPCC elements are carried out by the HPCC working group (chaired

by Prof Y. Matsuzaki, Science University of

Tokyo).

5. CONCLUTIONS

The Building Research Institute, Ministry of

Construction, Japan initiated a research program

on smart materials and structural systems in 1998 as a part of U.S.-Japan cooperative

research efforts. This paper described the

research items and plans of the research program

in Japan which have three research thrusts, (1) structural systems, (2) sensing and monitoring

technology, and (3) effector technology.

Although these three thrusts are common for the

activities planned to take place in Japan and in the U.S.A, the activities in Japan will have a primary emphasis on system applications and

proof-of-concept demonstrations while the

activities in the U.S.A. will focus on enabling

concepts and fimdamental aspects of the

technological elements. For the success of the

joint program, the exchange of researchers and

coordination and integration of cooperative

research are essential.

ACKNOLEDGEMENT

146
This work has been carried out under the USJapan cooperative structural research project on

Smart Structure Systems (Chairperson of

Japanese side: Prof. S. Otani, University of

Tokyo, Chairperson of U.S. side: Prof M. A.

Sozen, Purdue University). We would like to acknowledge the hard work and contribution of

structural systems, Prof. Y. Kitagawa, chairperson

of sub-committee on sensing and monitoring

technology, Prof T. Fujita, chairperson of subthe U.S. participants at the JTCC meeting in Tsukuba, 6-8 January, 2000. The outcome of

discussion at the meeting is fundamental to this paper. The authors also would like acknowledge Prof. A. Wada, chairperson of subcommittee on

committee on effector technology, and all members of the project for their useful advice and

suggestions.

147
Table 1 Smart Stmctural Systems for Aerospace Engineering and Building Engineering

Aerospace Engineering BuildingCDEngineeringo

Characteristics of

structure

Airplane is originally active and

adaptive.

Building is not required to be active or adaptive.

Airplane has a simple usage and works

as a single unit. A group of buildings form a social

unit having multiple usage. External

disturbance

and objective safety

Structure must be safe in daily usage and

disturbance.

Structure must be safe in rare events,

such as strong winds or an earthquake.

Constant maintenance is required. Free-maintenance is desirable.

Research needs Integrate smart functions into a structure

to achieve light-weight and highperformance.

Put smart functions to a structure to achieve objective performance such

as minimum life cycle cost. Typical example Active control system Health monitoring system

Table 2 Research Needs in Building Engineering

Category Research needs

Function Effective control of noise and vibration

Creation of large open space without column and wall

Design of highly irregular buildings

Flexibility in building usage

Extension of building life Disaster prevention Prevention of building or ground collapse

Rehabilitation of old structures Damage detection of hidden structural elements

Human safety at ultimate stage

Repair of building damage

Evaluation of seismic safety in urban environment

Public education toward disaster mitigation

Environment Protection of nature Control of environmental pollution

Control of industrial waste

Reduction of dust and noise during construction

Production Countermeasures against shortage of expert builders

Improvement of construction quality

High speed construction

Development of new material

148
Table 3 Examples of application of SMA in Japan

Properties of SMA Examples of application

Shape memory effect Control of room temperature using SMA sensors and actuators Door unlock system for earthquakes using SMA actuators Deflection control of structure using SMA wires

Stress control of pre-cast concrete members using SMA springs

Demolition of a part of concrete structures using SMA devices

Joints of steel pipes using SMA tubes

Framework of reinforced concrete curved membranes using SMA supports

Pseudo elasticity RC members with longitudinal bars of SMA

Isolation system of containers for vibration

Base isolation device with the combination of SMA, rubber and lead

Seismic energy absorbing device for historic structures or bridges

149
Beauty

Surrounding

Environment

Economy [Value of buildings^ Strengtl)

/

Earth! Environment

Function

Figure I Evaluation Elements of Building

o

o

o

HI^Rninnxfi/q o

o

o

' o

HiaHxinnxfs/Q o o o o O 1" ~oo

o

ca

c

3QQQ I 2000 I

Figure 2 A five-story steel frame model (unit:mm)

kN A~—

T:350

0 2 4 %

kN

} \

—

T:500

kN

% 0 —-J t

\^—

'

T.550

—I—

T:450

% 0 % 0

Fig. 3 Stress-strain relations of specimens under

deferent temperatures at shape memory treatment

(cf. test condition : in room temperature)

150
151
ISA Devices

Fie. 7 Active control of floor Fig. 8 Friction control for base isolation

deformation

embedded

Column or beam

Fig. 9 Sensing system with or without

cabling

Fig. 10 PVDF sensors for membrane

152
. 2 ) 2 )

m^rmit. (1) riiint^ii^^SatvX-f A(7)j5>t:tttiH^lf-f[ffi. (2) Ml^miil^liiigcOli^J-n. (3)

: ^DBB J

(3) ^ (i^ : mm Pt^ M^^±^^mmm^pmm

JE

public corporations: BCJ BCS HUDC

KKSK

private corporations: Kyowa Dengyo Co. (Sensing) NIhon Ultrasonic Testing Co.

(Sensing) NTT Facilities Co. (Sensing) Daldo Steel Co. (Effectors) Sho-bond Construction Co.

(Effectors) Kowa Chemical Industry Co. Konishi (Ettectors)

(Effectors)

BRI NSF

JTCC

NEDO

PWRI

Monbusho JSCA

universities KC

BRIC

Research Committee

Japan Side

TCC

(Prof. S Otani)

Smart

Structural Systems

(Pro< A Wada)

S«n«lng &

Monltonng

(Prof Y Kitagav/a)

Effector*

(Prof, T Fu|Ha)

WGs WGs WGs

Secretariat: BCJ

Research

• Committee - US Side

TCC

(Prof M. A. Sozen)

Co<iatxxal>on Joint Researcri 153

BCJ Building Center o( Japan BCS Building Contractors Soctety

BFll Building Research Inslitule BRIO Building Rosonrc*i Institirtc Community HUDC Housing 4 Urtjan Developfnenl Co. JSCA : Japan Structural Consultants Assoc KKSK KerKlliku Kenkyu Shinko Kyokai KC Ko/ai Club NE(X> . New Ert«fgy & tnduslnat TecTrrwlogy O^vaiopment Organization NSF US Natkxial Science Foundalbn PWRI Public WofVs Research Insiiluta

Nonlinear Rock Mechanics Analysis of Gravity Dams

under Combined Flood and Earthquake Loading

by

Luis A. de Bejar' and Robert L. Hall'

ABSTRACT

The two-dimensional nonlinear incremental

response of a gravity dam subjected to earthquake loading is outlined. The dam is assumed to stand on jointed rock and to be

impounding a full-reservoir upstream water

level. The material constitutive relations for

rock discontinuity are expressed in terms of an incremental plasticity model that fully

represents the coupling between joint normal

and shear deformations due to dilatancy. The

model also accounts for stiffness degradation

of joints upon wear and shear distortion, and

for the bulking/seating effects of separated

rock debris. The instantaneous center of

rigidity and the associated principal directions

of rigidity for the dam-rock system are located

dr.di;, ;;caily for each step of the load history,

and the displacement and force responses of

the rock joint elements are obtained

incrementally. Results are compared to those

from corresponding finite-element analyses.

KEYWORDS: Bottom-absorption

hydrodynamic force spectra; finite-element

modeling; flood, nonlinear earthquake

analysis; rock mechanics.

1. INTRODUCTION

In order to build the model in this

investigation, several recent studies dealing

with the behavior of rock discontinuities

under shear and normal loading were

examined. Saeb and Amadei (1992) developed

a tangent formulation for the deformability of

a rock joint that fiilly accounts for the coupling

between joint normal and shear responses due

to dilatancy. Plesha (1987) and Qiu et al (1993) have developed a macroscopic

constitutive law for initially closely mated

joints undergoing small displacements that

includes dilatancy, damage of surface

roughness, and cyclic sliding behavior in the

light of the theory of plasticity. This theoretical

model was experimentally verified for simple

laboratory experiments on both natural and

artificial rock joints (Huang et al. 1993). Mroz

and Gambanco (1996) developed an interface

constitutive model accounting for slip and

sliding effects. The microslip effects were

described by considering the interaction

between asperities modeled as spheres in contact by appealing to the generation of

progressive or reverse slip zones.

The constitutive law for rock discontinuities

applied in this investigation mirrors Plesha's

model (See de Bejar, L A. & S. Gamer 1998).

Associated-flow plasticity was adopted

because it results in symmetric stiffness

matrices that dramatically simplify the ensuing

numerical analysis procedures. Joint asperities are modeled as harmonic elements in regular

array, starting from the fresh-joint perfectly

mating configuration. Degradation is modeled

in terms of an exponential decay of inclination

with respect to the initial overall joint plane, as a function of the cumulative plastic work on

the asperities. Bulking/seating effects are built into the normality rule and subsequently into

the consistency condition of the theory.

Bulking effects may be disregarded for

monotonically increasing load but must be

included in the formulation for problems

involving reversal of frictional shear in the case

U.S. Army Engineer Research and Development Center 3909 Halls Ferry Road, Vicksburg, MS 39 1 80-6 1 99

www wes.army.mil

155
of earthquake loading.

2. MODELING THE DAM-FOUNDATION

ROCK SYSTEM

The dam-parent rock system is idealized as a

rigid body standing on a set of elasto-plastic

rock discontinuities (Figure 1). The surface of

contact of the rock discontinuities is idealized as sinusoidal (Figure 2). This assumption

allows easy calculation of the displaced

configuration as mating blocks friction slide

and climb on each other during dynamic

response. Each discontinuity resists normal

and tangential forces through stiffnesses that

evolve as the exciting load fluctuates in magnitude (Figure 3). The instantaneous

center of rigidity and the associated principal

directions of rigidity for the dam-rock system

are located analytically for each step of the

load history (Ayre 1938, de Bejar 1985). The

model includes the hydrodynamic pressure

field on the upstream face of the dam, effective

upon bottom absorption of seismic energy in the reservoir. The model also includes the

effects of a downstream pool and the resultant

uplift forces exerted by the pore water

pressure field within the rock joints on the

upper discontinuity surfaces. This last effect

varies continuously as the dam-foundation

rock system translates and rotates during the

response history.

3. ELASTO-PLASTIC CONSTITUTIVE

LAW FOR ROCK DISCONTINUITIES

In this model, stresses are assumed uniformly

distributed over each segmental rock joint;

however, there is no limit as to the number of

joints to be considered in a given problem. The

Coulomb dry friction slip criterion is taken as the slip surface to signal the onset of plastic

flow in rock discontinuities, and is given by

(Qiu et al. 1993):

^({/}.a) I/,- cos a Z^- sin a I* _

* M' ( sin a cos a ) - 0 ^

'

where {f}= <fl = vector of internal forces

on the rock joint, a= the instantaneous

tangential inclination of the point of contac

between lower and upper faces of a joint

asperity, and \x = static coefficient of dry

friction for the rock joint. The same function

F({f}, a) is specified as the plastic

deformation potential leading to associated

flow. The formulation of the theory of

plasticity leads to an explicit relation between

the increment of displacement and the

corresponding increment of internal force,

which may be written as:

[E]-{d,} ,F<Oor,F<0

[[£] {dg} , F-0 and dF'O

where [E] = Elastic stiffhess matrix (diagonal)

for the rock discontinuity, and

[E*] = Elasto-plastic stiffness matrix (fully

populated and symmetrical).

4. NUMERICAL ANALYSIS

The two-dimensional response of the gravity

dam standing on jointed rock is calculated

incrementally. At each step in the analysis, the incremental displacement response of the

system is obtained after adjusting the material

state (elastic/plastic) of all joints for

consistency with their corresponding states at the end of the step. If, during the analysis for a given step, lack of consistency is detected

between the initial and the final states of any

joint in the set, the analysis is repeated

altogether for the step, using half the current value of the time step.

The incremental response for a given step is obtained as the summation of partial

contributions obtained from successive

attempts to restore dynamic equilibrium under

the effective earthquake forces, which include

both the inertial and structural damping

components. At each attempt, the partial

response of the system is obtained using its current stiffness matrix at the beginning of the

step and the principal rigidity properties for the

system, obtained from the elements as if their

normal and tangential actions were uncoupled.

This process is graphically illustrated by the

156
flow diagram in Figure 4 (for the onedimensional case), in which ({f},{S}) and

({F},{A}) are the effective earthquake force

and displacement response pairs for the joints

and the system, respectively (Quantities for the

system are referred to the principal rigidity

coordinates attached to the center of rigidity).

As a simple example of application. Figure 5 shows an idealized dam-rock system under a comprehensive representation of the

equilibrating forces during earthquake

response. The hydrodynamic pressure field D

in this representation is effective, after bottom-absorption eflfects have been estimated

from previous studies (de Bejar, L.A. 2000).

Figure 6 shows a typical set of hydrodynamic

force spectra normalized with respect to the

corresponding scaled hydrostatic resultant

force on the upstream face of the dam, for

various degrees of elasticity of the underlying

supporting rock (represented by the coefficient

of reflection a), and developed for the San

Fernando earthquake of Feb. 9, 1971. Figure

7 shows the distribution of the shell major

principal stresses of the system standing on

nonlinear gap elements simulating the rock

discontinuities. An extreme-load scenario of

simultaneous flood-earthquake excitation has

been assumed to emphasize the detrimental

effects on the system under such conditions.

The results of the analysis using the rigid-body

engineering model are contrasted with those

from the finite-element formulation for

comparison.

5. SUMMARY

An engineering model for the nonlinear

behavior of rock discontinuities under concrete

gravity dams subjected simultaneously to flood and earthquake has been outlined. The

material representation for rock joints had

been previously verified in closely controlled

laboratory experiments. The resulting

computer capability is intended to examine

effectively the progression of damage

associated with specific states for concrete

gravity dams during earthquake response. In

particular, the model will assist in the

definition of limit states and the construction

of fragility curves for comprehensive risk analyses (de Bejar, L A 1999)

6. ACKNOWLEDGMENTS

This investigation was conducted at the U S

Army Engineer Research and Development

Center, Waterways Experiment Station

through the Earthquake Engineering and the

Risk Analysis for Dam Safety Research

Programs, both sponsored by Headquarters,

U.S. Army Corps of Engineers. The authors

gratefijlly acknowledge permission from the

Chief of Engineers to publish the information

in this paper.

7. REFERENCES

1. Ayre, R.S. 1938. Interconnection of

translational and torsional vibrations in buildings. Bull. Seismol. Soc. Amer. U2.

28:89-130.

2. de Bejar, L A. 1985. Linear and nonlinear

analyses of building systems subjected to multicomponent earthquake excitations. Ph.D.

Thesis at Cornell University. Ithaca, NY.

3. de Bejar, L A. & S Garner 1998. Nonlinear

rock mechanics analysis of gravity dams under

flood loading. Proceedings Int. Symp. New

Trends and Guidelines on Dam Safety.

Barcelona, Spain.

4. de Bejar, L A. 1999. Fragility analysis for

concrete gravity dam safety under flood.

Proceedings of ASDSO Dam Safety

Conference. St. Louis, MO.

5. de Bejar, L A. 2000. Reservoir-bottom

effects on hydrodynamic loads for concrete

gravity dams subjected to earthquake.

Proceedings of ASCE J 4th Engineering

Mechanics Conference. Austin, TX.

6. Huang, X., B.C Haimson, M E. Plesha &

X. Qiu 1993. An investigation of the

mechanics of rock joints—Part I. Laboratory

investigation. Int. J. Rock Mech. Min. Sci. &

Geomech. Abstr. 30(3):257-269.

7. Mroz, Z. & G. Giambanco 1996. An

157
interface model for analysis of deformation

behavior of discontinuities, hit. J. Numer.

Anal. Methods Geomech. 20: 1-33.

8. Plesha, M E. 1987. Constitutive models for

rock discontinuities with dilatancy and surface

degradation. Int. J. Numer. Anal. Methods

Geomech. 11:345-362.

9. Saeb, S. & B Amadei 1992. Modeling rock

joints under shear and normal loading. Int. J. Rock Mech. Min.Sci. & Geomech. Abstr

29(3):267-278.

10. Qiu, X., M.E. Plesha, X. Huang & B.C.

Haimson 1993. An investigation of the

mechanics of rock joints—Part II. Analytical

investigation. Int. J. Rock Mech. Min. Sci. &

Geomech. Abstr. 30(3):27 1-287.

158
FIGURE 1 . Analytical model for dam-rock-discontinuities system

159
FIGURE 2. Idealization of asperity contact in rock discontinuity friction sliding

160
B

FIGURE 3. Analytical model for the dam-system components under earthquake

161
{A} = {0}

(6,} = {0}

{AP}

Effect of {f;} = {0}

{AP}""'^'= (AP}

NXT STEP

{A^^} =[K]-'- {AP""'"')

{A} -{A} +{A^"^}

KINEMATICS( {A^"^} )

{^} = [E>]-{5i}

Effect of{f;}

(AP}""''^= {AP} + Effect of {fj}

O

FIGURE 4. Global flow of numerical iterations within a given load step

162
FIGURE 5. Two-dimensional system of simultaneous gravity, hydraulic, and seismic

forces on dam-rock-discontinuities system

163
FIGURE 6. Hydrodynamic force spectrum for the San Fernando earthquake as

recorded at Whittier Narrows Dam, 1971 (S53W component)

164
0

Output Set: INCH 1, LOAD= 1.000E+00

Deformed(0.121): TOTAL TRANSLATION

Contour: SHELL MAJOR-PRINCIPAL-1

FIGURE 7. Typical 2-D dam-rock system standing on nonlinear rock joints.

Loading: Gravity + 5m overtopping + earthquake + uplift + sediments

165

OUTLINE AND PRE-ANALYTICAL STUDY ON DAMAGE DETECTION TESTS OF FIVESTORY STEEL FRAME WITH SIMULATED DAMAGES

by

TESHIGAWARA Masaomi'^, KAMINOSONO Takashi'^ INUKAI Mizuo'^ ISODA Hiroshi'\

MORITA Koichi^^ and NOGUCHI Kazuya'>

ABSTRACT

This paper presents outline and pre-analytical study on damage detection tests of five-story steel frame

with simulated damages. Fiber brag grating (FBG)

sensors, accelerometers and strain gauges will be

installed in this frame. From the results of preanalytical study, we can estimate which story is damaged from the change of natural period or mode shape to some extent. Damage identification

method using modal analysis is shown here. Damage detection methods from local information will be also examined, and various

identification methods will be also applied.

KEY WORDS : Smart Strucmre

Health Monitoring

Damage Detection

Identification Techniques

1.INTRODUCTION

If we need to reduce life cycle costs of a building

from construction to maintenance, it is very

effective to monitor structural health of a building.

However, the research on health monitoring of

buildings is rare, and health monitoring technique

of a buildmg is not still establishing. We have a plan of damage detection tests using smart

sensors, as fiber optics, and this is a part of USJapan cooperative research project on smart

structural systems. We show here outline of tests and results of pre-analysis.

2.0UTLINE OF TEST FRAME AND

EXPERIMENTS

Test frame is five-story steel structure shown as

figure 1 . Floor height is Im, total height is 5m and

floor plan is 2m*3m. The cross section of

members is following.

column : H148*100*6/9(SS400)

beam : H148*100*6/9(SS400)

stud : H100*50*5/7(SS400)

We put a weight of 2 ton on each story.

There are two test cases, tests with studs shown as figure 1.1) and tests without stud shovm as figure

1.2). Braces can be installed, when studs are removed. Natural periods are almost 0.3 - 0.4

seconds.

Test frame will be set up on shaking table, and

excitation tests using a vibration exciter will be

also carried out, if we need to excite higher

modes. Sensors will be put on each floor and

1) Building Research Institute, Ministry

of Construction, Tsukuba-shi, Ibarakiken, 305-0802, Japan

installed at beam edges. Fiber brag grating (FBG)

sensors, accelerometers and sfrain gauges will be

installed. In addition, ultrasonic measurements

will be carried out. Settings of simulated damages

are as fellow. +Studs or braces will be extracted only from one

story.

+Part of flanges of beams or columns will be cut

off. (See figure 2)

+Bolts of beams or columns will be loosened.

Etc. The following tests from i) to vii) will be carried out in order to the number.

i) Normal sound test structure with studs

ii) Removing studs from only one story

iii) Normal sound test structure without studs

iv) Loosening bolts of beams of test structure without studs

v) Exchanging sound beam for damaged beam

vi) Exfracting braces from only one story

vii) Cutting of part of column

3.PRE-ANALYSIS

(l)Change of modal property by setting

damage

167
We examine how natural periods and modal

shapes change when damages are set at part of

beams or columns. Some cases are examined, and

all results tend to be same. Here, only the case in which damages are set at beam edge of test structure without stud is explained. Damages at beam are assumed as pin in analysis.

Table 1 shows that natural periods of test structure, in which damages at beam edge are assumed, are compared. First natural periods

greatly change, when damages at lower story are assumed. Higher (third) natural periods greatly

change, when damages at upper story are assumed. Figiu-e 3 shows similar results of mode

shapes. First mode shape also greatly changes,

when damages are assumed at lower story. Higher

(third) mode shapes also greatly change, when

damages are assumed at upper story. From these

results, it is effective to examine higher mode

when there seem to be damages at upper story. We

can estimate which story is damaged from the

change of natural period or mode shape to some

extent.

(2)Damage Identification by use of change of

modal property

We make simulation waves calculated as results of same models used in (1) with white noise for

input. Identification technique [1] using the

change of flexibility is applied to this simulation waves. Damaged location is estimated by

comparison of sound modal properties and

damaged modal properties. We show here results

of the case setting damage at beam edge of 1, 3, 5

story in figure 4-6. Each figure shows the flame

in case of figure 1 2). Y direction is height

direction and dark line shows that damage index is large. That is to say, if the color of lines is dark,

there should be damage. In figure 4 (assumed

damages at lower story), the color of lower story

is dark. In figure 5 (assumed damages at intermediate story), the color of intermediate story

is dark. In figure 6 (assumed damages at upper

story), the color of upper story is dark. From these

figures, damage identification seem to be difficult when damage at upper story is assumed. And we

can estimate which story is damaged when

damages exist at lower story.

4. Conclusions

Outline of damage detection tests and results of

pre-analysis are reported. Damage identification

method using modal analysis is shown here. Damage detection methods from local information will be also examined, and various

identification methods will be also applied.

[Reference]

[1] "Damage Identification and Health Monitoring

of Structural and Mechanical Systems from

Changes in Their Vibration Characteristics" : A

Literature Review, Los Alamos National

Laboratory Report, LA-13070-MS, 1996

Table 1 : Comparison of Natural Periods sec]

1st 2nd 3rd

Sound structure 0.366 0.120 0.0710

Damaged beam at 5^^ story 0.371 0.131 0.0786

Damaged beam at 4* story 0.385 0.141 0.0745

Damaged beam at S""^ story 0.409 0.132 0.0729

Damaged beam at 2"<^ story 0.431 0.120 0.0768

Damaged beam at 1^* story 0.437 0.129 0.0716

I

168 i

It
I 169
no damage no damage no damage - - damage at IF beam ~ ' damage at IF beam - - damage at IF beam

daroge at 3F beam — - damage at 3F beam ~ * damage at 3F beam

damage at 5F beam damage at 5F beam damage at 5F beam

Corparison of 1st mode shapes Corparison of 2nd mode shapes Corparison of 3rd mode shapes

Figure 3 : Comparison of Mode Shape without damages and Mode Shapes with damages

Ftx±)iliry M ethod -FJ&xfcitSy Change

Flsx±iil3^ M «thod -Flexbilty Change

170
Ficnire^ • Tn Casp of Slettinp Damapes at V .Stnrv

171
Ii
The influence of high confining stress on the cyclic behavior of saturated sand

by

R. Scott Steedman', Richard H. Ledbettei^ and Mary E. Hynes^

ABSTRACT

This paper describes the findings of an experimental study supported by the U.S. Army

Centrifuge Research Center and Engineer

Earthquake Engineering Research Program

(EQEN) into the behavior of saturated sands

under high initial effective confining stresses subjected to strong ground shaking. The

research was conducted using the Army

Centrifuge at the U.S. Army Engineering

Research and Development Center (ERDC),

located in Vicksburg MS, formerly known as the

Waterways Experiment Station (WES). The

centrifuge studies have shown that the

generation of excess pore pressure is capped at a

level below 100 percent for vertical effective

confining stresses exceeding around 3 atmospheres (atm, or 300 KPa). This limit

reduces at higher confining stresses. If verified,

the potential benefits from this finding for the

design of remediation works for large earth dams could be substantial. The paper describes

the equipment used for the experiments, the

research program, and presents the initial results,

contrasting the development of excess pore

pressure at low confining stress with that at high

confining stress. Possible consequences for a hypothetical dam are discussed.

KEYWORDS: centriftige; dams; ERDC;

liquefaction; modeling; remediation

1.0 INTRODUCTION

The current state-of-practice for the evaluation

of liquefaction potential and for remediation

design and analysis depends on empirical

correlations of in-situ measurements of strength

versus field experience of liquefaction at shallow depth and laboratory data of the

behavior of confined elements under cyclic

loading. (Liquefaction is defined here to mean

the development of pore pressure equal to 100%

of the initial vertical effective stress.) This

approach is known as the "simplified

procedure". Opinions vary as to the maximum

depth in the field at which liquefaction has been

observed, but there is no established field evidence from historic earthquakes of

liquefaction at depths greater than a few tens of

meters. The NCEER Workshop in 1996 on the

Evaluation of Liquefaction Resistance of Soils

noted that the simplified procedure was

developed from evaluations of field observations

and field and laboratory test data, Youd and

Idriss (1997). The report notes, "These data were

collected mostly from sites ... at shallow depths

(less than 15m). The original procedure was

verified for and is applicable only to these site conditions".

Hence, in design practice the assessment of

liquefaction under high initial effective

confining stress, such as might relate to the

foundations of large earth dams, is based on the

extrapolation of observed behavior and

correlations at shallow depths. In practice, the

behavior of saturated soil under these conditions

is not well understood. Based on the results of

laboratory tests, researchers have postulated that there is a reduction in the liquefaction resistance

of such soil compared to shallow depths. This

reduction is accounted for in standard

approaches by a ratio known as K^, a "correction

factor" developed by Seed (1983) in the

simplified procedure. This strength ratio is postulated to reduce with increasing initial effective confining stress, which has potentially

' Steedman & Associates Ltd, 25 Eldon Square, Reading RGl 4DP, UK, Robert.S, SteediTian@erdc.usace.army.mil

^ US Army ERDC Waterways Experiment Station, Vicksburg, MS 39180, Mary, E.Hynes@erdc.usace.army.mil

173
large impact on the extent and method (hence

cost) of remedial construction required to assure adequate seismic performance of large dams.

For example, it may reduce the cyclic shear

stress ratio predicted to cause liquefaction in a

soil layer under a typical large dam (of the order

of 30m high) to about 50% of its value in the

absence of the dam. The predicted deformations

and resistance and the remedial strength required

are a direct flinction of the reduced shear

strength. The effects of the strength

reduction is driving the majority of seismic dam

safety concerns and needs for remediation,

which may result in costs in the billions of

dollars in North America alone in the coming

years.

The factor quantifies the curvature in the

cyclic shear strength envelope (cyclic shear

stress required to cause liquefaction versus confining stress) for a soil as observed in laboratory tests on discrete specimens. Although some curvature may be expected, such large

reductions in cyclic shear strength ratios are

counter-intuitive. It is generally accepted that increased confining stress should broadly

improve the capacity of a soil to resist applied

loads, not reduce it. Clearly the volume of soil which is required to be treated and the difficulty

and expense of that treatment are highly

dependent on an accurate assessment of the

potential for and consequences of liquefaction.

There is therefore strong motivation for owners

of large dams to investigate the behavior of

saturated sands subject to strong ground shaking

under high initial effective confining stresses. In the absence of field data, the use of a centrifuge was considered to be the only practical option to

realistically represent a deep soil deposit

subjected to earthquake shaking. The studies to date have focused on level ground initial stress conditions in a two-layer (dense over loose)

deposit of clean, fine Nevada sand. Examination

of more complex stratigraphy and sloping

ground stress conditions is planned for future

studies.

2.0 THE ERDC CENTRIFUGE,

EARTHQUAKE ACTUATOR AND ESB

SPECIMEN CONTAINER

The design specification for the ERDC

centrifuge followed a review of the Army's

research needs and a study of the available

academic facilities, Ledbetter (1991). Many of

the field problems with which the U.S. Army

Corps of Engineers is concerned are physically

large, such as earth dams, locks and river control

structures, environmental problems and military

research. It was determined that a new facility

with a high payload capacity and high g

capability was required to meet future Army

needs. A large beam centrifuge was

commissioned, termed the Acutronic 684, based on the French designed Acutronic 661, 665 and

680 series of geotechnical centrifuges, Ledbetter

et al. (1994a). The capacity of the ERDC

centrifuge is a payload of 8 tonnes at up to 143

g, reducing to 2 tonnes at 350 g, with a platform

area of 1 .3 m square. This high capacity enables

field problems of the order of up to 300m in breadth, 300m in depth, and 1000m in length to be simulated under a wide variety of loading

conditions. The facility is now fully operational,

and is equipped with a large range of equipment

and appurtenances. Figure 1.

The research approach for this high confining

stress liquefaction study uses the large capacity

of the WES centrifuge to investigate the

generation of excess pore pressure and

liquefaction under conditions which much more

closely resemble those at depth in the field,

Ledbetter et al. (1999). To do this required the

design and construction of a large dynamic

actuator.

2.1 Shaker design

A mechanical design was adopted for the

earthquake actuator for the ERDC centrifuge,

based closely on a smaller version designed for Cambridge University, England. Complex

servo-hydraulic shaking systems have been

developed and are operational on several

centrifuges around the world, but these are limited in the g level to which they may operate.

174
and are considerably more expensive to build.

The research by ERDC into the basic behavior

of liquefying soils and comparison with present

design methods required a base motion

comprising a series of uniform cycles of base

shaking, of variable duration and uniform

frequency. A rotating mechanical system is ideally suited for this purpose. The large

carrying capacity of the centrifuge meant that the shaker itself could act as the reaction mass

for the specimen, minimizing the vibrations

which would be otherwise transmitted to the

centrifuge itself, and the model container could

be much deeper and longer than usual.

The principle of the operation of the Mk I shaker was based on the use of stored energy to drive

the specimen back and forth. Flywheels are incorporated in the shaker mechanism to store the energy in advance of the shaking event in the

form of stored angular momentum. A system of

linkages and eccentrics transferred the stored

energy of the fly wheels to the shaking platform

and thence into the soil specimen, Figure 2. A

hydraulic or electrical motor drove the flywheels

up to full speed, and then, on a signal, a high

speed clutch grabbed the oscillating shaft and

transferred energy into the model until another

signal released it again. Clearly the frequency of

the oscillation was directly proportional to the

speed of the motor (and flywheels). The

amplitude was controlled by the arrangement of

the eccentrics; three displacement amplitudes for the platform were available (+/- 0.49mm, +/-

1.47mm and +/-4.41mm).

The shaker was designed (structurally) for operation up to 150g, at which the maximum

load capacity of the shaking platform is reached

(75 tonnes). The design maximum lateral force

which the mechanism could exert on the shaking

platform was 30 tonnes, and the maximum

frequency at which the shaker could be safely

operated was 150 Hz (eg. I Hz prototype at 150g, or 3Hz prototype at 50g). (The Mk I shaker has recently been upgraded.)

2.2 Model containment

The specimen is built within a hollow

rectangular model container, termed an

equivalent shear team (ESB) container,

comprising a series of eleven aluminium alloy

rings stacked one above the other, and separated

by an elastic medium, Figure 3. Several of these

chambers have been constructed, and extensive

dynamic analysis and testing has been carried out to determine their dynamic response

characteristics, Butler (1999). The model

container has internal dimensions of 627mm

deep by 3 1 5mm wide by 796mm long. Each of

the eleven aluminium alloy rings is 50mm high.

The rings are not stiff enough along their long

dimension to support the outward pressure from

the soil inside under high g, but they are supported by the massive reaction walls of the

shaker unit itself A rubber sheet separates the

rings from the steel walls on either side. This

concept has the added advantage of raising the center of gravity of the reaction mass in line with the center of gravity of the specimen, thus

minimizing eccentric forces that may lead to rocking.

Thin metal sheets, termed shear sheets, are positioned on the interior end walls of the

chamber and fixed securely to its base. The

shear sheets accommodate the complementary

shear force generated by the horizontal shaking

within the specimen and transmit that force to the base of the container, Figure 4. This

improves the uniformity of the stress field at each elevation along the model, reducing the

tendency for the chamber to 'rock'.

The ESB concept is to create an equivalent shear

beam with an average stiffness comparable to the stiffness of the soil specimen. Expressed

rigorously, the concept is more accurately

defined as achieving a dynamic response that does not significantly influence the behavior of

the soil specimen inside. In certain classes of

experiment, where the zone of interest is limited

to the central region of the chamber, it may be

expected that the stiffness of the soil (at least near the end walls) would not reduce

significantly during shaking due to excess pore

pressure rise. In this case the stiffness of the

chamber may be designed accordingly, perhaps

considering a shear modulus appropriate to the

level of dynamic strain expected in the soil freefield at mid-depth. For other experiments

175
involving the liquefaction of large volumes of

soil inside the container, the stiffness changes

(and hence dynamic response) will change

dramatically throughout the base shaking. A

stiff chamber may lead to undesirable effects, as noted by Peiris (1999) who observed that liquefaction in a loose saturated sand model did not occur near the stiff end walls of the chamber.

A chamber with no stiffness simply adds mass to the soil specimen, again changing its dynamic

response. This poses a particular challenge for the design of an ESB container.

The ERDC ESB used in these experiments was

assembled using a urethane adhesive sealant

(commonly used as a windshield sealant for cars) between the aluminum alloy rings. This

material has good elastic properties (exhibiting

only minimal hysteresis under cyclic loading)

and bonded well to the metal and to itself. The

ESB has a relatively low shear stiffness of 441

KN/m^ (shear stiffness of the full stacked ring

assembly) and mass of 229 kg, with a first mode

at 16Hz, and second, third, and fourth modes at 46, 87, and 1 16Hz respectively (Butler 1999). A

typical saturated specimen at 50g in the ERDC

ESB will have a theoretical natural frequency of

around 84Hz, based on an average small strain shear modulus of 96 MN/m^.

2.3 Dynamic response of the model container

and specimen

In his doctoral thesis, Butler (1999) completed a thorough theoretical and experimental analysis

of the dynamic response of the coupled soil- container system. At high g the soil and

container act as a coupled system, where the

lower stiffness of the container reduces the

natural frequency (slightly) of the combined

system compared to the soil column alone.

However, provided the driving frequency is low

relative to the natural frequency of the coupled

system, Butler (1999) demonstrates that the

displacement response of the system is unaffected compared to the soil acting

independently, an ideal situation.

For higher driving frequencies, Butler (1999)

concludes that it would be necessary to reconsider the elastic stiffness of the ESB

container, and to tune the container to ensure

that even with the expected level of degradation

in the soil specimen, the coupled system did not

deviate significantly from the condition of the

soil column alone. This may be possible by

adding mass to the rings of an initially stiffer ESB to reduce its first mode to the desired level,

Butler (1999).

The liquefaction of a level sand bed has

previously been the subject of other research.

Experiments were conducted at many centrifuge

centers under the VELACS project,

Arulanandan and Scott (1994). The objective of

the ERDC study is to investigate the onset of

liquefaction under much higher initial effective

overburden stresses.

3.0 RESEARCH PROGRAM

Table 1 summarises the experiments conducted

during 1998 and 1999. The models are grouped

in series, where each series corresponds to a

different target range of vertical effective

overburden stress in the loose layer. In all cases, the bottom 160mm of the specimen was around

50% Relative Density (RD) and the upper

portion was around 75% RD. All models were

shaken at 50g. Some models were

overconsolidated by a factor of 2.5 prior to shaking (achieved by running the centrifuge up

to 125g). A large series of experiments have

been conducted with a range of overburden

depths to provide repeatability and redundancy

in the instrumentation and data records. All the

models were built using Nevada sand and tested

at 50 g. The experiments are summarized in Table 2.

The Nevada sand used in the models was

characterized by standard laboratory tests to determine parameters such as dry density and

gradation. Table 3 presents key material

parameters for this sand.

The pore fluid comprised a mixture of glycerine

and water, 80% by weight for experiments

conducted at 50g. Measurements of the viscosity

of glycerine-water mixes at a range of

temperatures and proportions show that the

viscosity is sensitive to both parameters.

176
The density of a glycerine-water mix was

calculated from:

Pm = p;.(mg + mw)/( mg + mw)

where is the density of the mix, Pg is the

density of glycerine, mg is the mass of glycerine,

and m^ is the mass of water. Table 4. summarizes the properties of the glycerine-water

solution used as the pore fluid.

The models were poured dry from a hopper and

saturated under vacuum, or slowly under

gravity. Instrumentation was placed in the

model as it was being constructed.

The typical arrangement of the model specimens

for the different test series is shown in Figure 5. Instrumentation was positioned through the

depth of the models, and comprised pore

pressure transducers and accelerometers. The

location of the instrumentation for Model 5b is shown in Figure 6.

4.0 AMPLIFICATION OF BASE INPUT

MOTION

As the base input motion propagates through the

height of the specimen, it may be amplified or attenuated depending on the dynamic properties

of the soil column and the frequency content of

the input motion. This is similar to conditions in the field, where elastic wave energy is modulated as it passes through different soil deposits. Examination of the time histories of

acceleration at different depths in a deep soil column such as Model 5b, Figure 6, shows that

in the absence of significant degradation caused

by large strains or excess pore pressures, the

amplification factor is around 1, Figure 7. A

similar response was also found for shallow

specimens. In specimens where large excess pore pressures started to develop through the soil column, however, the characteristics of the

motion propagating upwards tended to change

dramatically. In the extreme, the lateral accelerations of the surface becomes isolated

from the base input, as may be seen in Figure 8

which shows the acceleration at different depths

in Model 2f

5.0 DEVELOPMENT OF EXCESS PORE

PRESSURES AND LIQUEFACTION AT

SHALLOW DEPTHS

The target initial effective vertical stress in Model 2f was 1 tsf (100 KPa) at mid-depth in the loose layer (Table 2). The specimen was

overconsolidated by accelerating the centrifuge

to 125g prior to subjecting it to shaking motion

at 50g. Figure 9 shows how the excess pore

pressures at different depths developed with

time. It is clear that within a few cycles of

shaking, the loose sand layer has fully liquefied.

Comparison between the acceleration time

histories in Figure 8 and the excess pore

pressures in Figure 9 for the loose and dense

layers shows the correlation between transmitted

motion and level of excess pore pressure. The

upper section (near surface) of the dense layer

liquefies within 0.2 seconds (0.2 x 50 = 10

seconds field equivalent). The percentage of

excess pore pressure is calculated by dividing

the measured fluid pressure by the calculated

initial overburden stress. The actual overburden

stress may vary over time due to small

movements of the transducer in the liquefied

ground; this has a larger effect on the accuracy

of measurement at shallower locations, such as PPT 38.

It is interesting to note that despite the difference

in relative density between the dense and loose

layers, the soil column behaves more like a

single unit than as two layers. This is because

each increment of excess pore pressure at depth

immediately causes a fluid pressure, which

affects the whole column through the volumetric

compressibility of the pore fluid. This is considerably lower than the shear

compressibility of the soil skeleton, which is determining the speed of transmission of the

elastic shear wave upwards.

6.0 GENERATION OF EXCESS PORE

PRESSURES AT HIGH CONFINING STRESS

At high initial effective confining stress pore

pressures are also observed to rise strongly

177
during the initial cycles of shaking. After some

time, this steady rise is arrested, and then excess pore pressure is capped at a vahie less than the

full 100% of the initial vertical effective stress seen in shallow soil columns. Figure 10 shows a

set of data from Model 4d, a deep soil specimen

with an initial effective vertical stress of around

500 KPa (4.7 tsf) at mid-depth in the loose layer.

A similar picture was found at even higher

initial vertical effective stress. Figure 1 1 shows

the development of excess pore pressure under a

vertical effective stress in the middle of the

loose layer of 1037 KPa (9.7 tsf). In this figure,

the actual fluid pressure has been plotted in KPa,

to be compared with the initial vertical effective

stress. The time history of base shaking is shown below the pore pressure record in two

forms. The first trace shows the measured

record, which includes a degree of high

frequency noise. In the second, lower trace, this signal has been processed with a Butterworth

band pass filter, 250 and 10 Hz, to extract the

main shaking frequency. From this record, it is clear that the amplitude of the shaking motion is almost constant throughout the record, and

particularly that the amplitude remains at or near

its peak level throughout the second half of the

shaking episode, when the excess pore pressure

has been capped (after about 0.8 seconds of

shaking, or 40 seconds field equivalent). In this case, the excess pore pressure is capped at around 50% of the value required to reach

'initial liquefaction'.

The amplitude of cyclic shaking was easily

sufficient to generate excess pore pressures at a strong rate during the first half of the event, and

this can be seen in the top trace of Figure 1 1

.

Indeed, after around 0.4 seconds of shaking the

rate of excess pore pressure build-up increases,

and this may be related to a moderate increase in the amplitude of base shaking at the same time.

Between around 0.4 seconds and 0.7 seconds of

shaking the rate of increase of excess pore

pressure is around 350/0.3 ~ 1000 KPa/second

(or 20 KPa/second field equivalent). The driving

frequency of input motion is around 50 Hz (1 Hz

field equivalent). During the same period, the

amplitude of shaking is around +/- 15% g, which

corresponds to a cyclic displacement of +/- 0.76

mm (38 mm field equivalent) at that depth.

'

This pattern of behavior was found in all specimens at initial vertical effective confining

stresses of around 250 KPa or greater. Although

not a rigorous comparison, as there are a number

of variables (including amplitude) between the

different experiments. Figure 12 shows the

normalised results from a range of initial conditions. There is a general trend with

increasing initial vertical effective stress, towards the capping of excess pore pressure at a

level lower than the expected 100%.

All of the records of excess pore pressure from

all the model specimens were re-examined to determine the residual pore pressure at the level of capping. Although the amplitudes of shaking

and the densities of the sand varied (the data-set

includes both 'dense' and 'loose' sand layers as described above) a clear correlation was found

between the maximum level of excess pore

pressure and the initial vertical effective stress, as shown in Figure 13. This figure includes 125

data points from first and succeeding

earthquakes, normally and overconsolidated

specimens, and dense and loose layers.

Figure 14 shows the data from first earthquakes

only, broken out by normally consolidated and

overconsolidated (including dense and loose

results). Again, there is a strong correlation

between the maximum level of excess pore

pressure generated during the shaking and the

initial vertical effective stress. In this case, the

overconsolidated data is seen to lie to the left of

the normally consolidated data, indicating that the level of excess pore pressure may reduce for increased overconsolidation. This is consistent

with experience.

On the right hand side of Figures 13 and 14, an approximate depth scale has been given for guidance, expressed in feet. Clearly the

calculation of equivalent field depth depends on many assumptions, but it is worth noting that the

data appear to show that whilst liquefaction may

readily occur within the upper 40 feet or so of a

saturated sand deposit, it becomes increasingly

178
unlikely at greater depths. This is consistent

with field observation.

Research work is continuing with a view to full verification of these findings and the preparation

of design guidance. Further experiments will address the effects of amplitude of shaking,

sloping ground and density.

7.0 CONSEQUENCES FOR A

HYPOTHETICAL LARGE EARTH DAM

To illustrate the potentially highly significant

consequences of capped pore pressures decreasing with depth, a hypothetical earth dam

of the form outlined in Figure 15 is assessed.

The estimated deformation response is calculated for earthquake induced potential pore

pressures based on standard approaches. Seed

and Harder ( 1 990), Youd and Idriss ( 1 997). and on limiting pore pressure generation.

Deformations are estimated following the

procedures of Ledbetter and Finn (1993), Finn

et. al. (1995), and Ledbetter et. al. (1994b), with

the use of the non-linear dynamic effective stress analysis program TARA, Finn el. al. (1986), and

Finn and Yogendrakumar (1989).

The dam is located across stream channel

deposits with assumed profile. Figure 15, as: (a)

foundation older stiffer alluvium below

elevations 151m (u/s) - 163 m (d/s), (b) younger

less stiff alluvium above the older alluvium, and

(c) roller compacted zoned earth dam founded on the young alluvium. Crest level and Pool

level are 211m and 199m respectively. Standard

Penetration (Ni)6o values range from 5 to 15 and

shear-wave-velocity (Vs ) from 180-450 m/s in the young alluvium freefield upstream and

downstream. Beneath the dam, the young

alluvium downstream (N|)6o values range from

30-40 and Vs from 425-700 m/s and upstream

(Ni)6o values range from 6-20 with Vs from 1 50-

350 m/s. Ground motions are assumed to be

from a magnitude 7.75 earthquake occurring at a distance from the dam about 50 miles with a peak acceleration of 0.22g at the dam.

7.1 Maximum potential residual excess pore

pressure

Seed's simplified procedures (Seed and Harder,

1990) were coupled with non-linear dynamic

effective stress finite element analysis to yield

"field observation and laboratory based"

estimates for earthquake induced pore pressures. These pore pressures are considered the

"maximum potential residual excess pore

pressures" that that could be generated by the

design earthquake, based on historic field earthquake response. Unless the soils have low

permeability and/or are bounded by low

permeability layers that prevent the dissipation

of excess porewater pressures, earthquakegenerated porewater pressures should be less than the maximum potential.

For this example dam, Figure 1 5 shows the

maximum potential residual excess porewater

pressures (shaded zones) that can be generated

by the design earthquake. The young alluvium

can generate pore pressures (a) downstream of

the downstream toe but nothing of significance

beneath the downstream shell and (b) upstream

of the upstream toe and beneath the upstream

shell. The lighter shaded upstream zone extends

completely beneath the darker shaded zone. Pore pressures in these lighter shaded zones are between 80 - 100 % of the initial vertical

effective stress with the majority between 90 and

100%.

The research findings expressed in this paper are

for level ground conditions. Continuing

research will be addressing the effects of

residual shear stresses such as induced by the

geometry of a dam. Until the geometry effects are known, this example applies the pore

pressures trend shown in Figure 14 for the

normally consolidated case as the cap or

limitation. In this example, the zone of soil that

will be effected and have reduced pore pressure

potential is the darker shaded zone beneath the

upstream shell; quite naturally, this is a high

confining stress zone and where leads to a

strength reduction due to the high confinement.

This zone is exactly where strength increase or

less strength degradation is of prime importance

179
to the dam behavior under earthquake loading.

The pore pressure potential in the darker shaded zone is reduced from 80 - 1 00 % to 50 -75%

of the vertical effective stress.

7.2 Deformation response of the dam

Figures 16 and 17 illustrate the effects or differences in earthquake induced deformations

and strains caused by the applied pore pressure

limits. For the deformation stage shown in the

figures, the only elements that appear shaded are those significantly straining and participating.

Figure 16 shows the deformations at effective

strength reduction to 72 % of the initial caused

by pore pressure increase. The dam will

continue to deform as pore pressure builds to the

maximum potentials and residual strength is reached beneath the dam in the foundation,

which is around 90 % reduction of initial strength. Unless remediated, severe earthquake

damage to the dam is possible accompanied by

overtopping and piping through the core

resulting in complete failure.

Figure 17 illustrates the benefits achieved by

applying the limiting pore pressure with depth

trend. The deformation stage in Figure 1 7 is the same as that in Figure 16. As shown strains,

deformations and the active elements are

significantly reduced. The dam continues to deform as pore pressures increase to their

maximum potentials and residual strengths;

however, the final stages of deformation are considerably less than those under the conditions

in Figure 16.

Immediately obvious from comparison of

Figures 16 and 17 is that more strength is available to resist deformation movements and

to buttress and assist remedial treatments. The

availability of more strength translates to smaller

and less remediation and lower costs. The stage

of progressive strength reduction illustrated in both Figures 16 and 17 was chosen to best

exemplify the impact of limiting pore pressure

criteria on dams with liquefiable soils in high

initial effective confining stress fields.

The implication of the present study is that although soil at depth may soften and degrade

under earthquake induced excess pore pressure

behavior, the reduction in strength is limited and

liquefaction under high confining stresses may

not be the hazard that it is presently assumed to be.

8.0 CONCLUSIONS

1 . A large data-set of the behavior of loose

saturated sands under high initial effective confining stresses and subject

to earthquake-like shaking has been

collected during an extensive

experimental program on the ERDC

Centrifuge, Vicksburg MS.

2. This data has shown that, at moderate

amplitudes of excitation, the maximum

level of excess pore pressure

development is capped at high initial effective confining stress and does not

reach a level sufficient to cause 'initial liquefaction' (defined as 100% of the

initial vertical effective stress).

3. These findings are currently being

verified and will then be used to develop

appropriate design guidance.

4. High pore pressures and the potential for

liquefaction beneath dams may not be

the hazard that it is currently perceived

to be.

5. The implication for the assessment of

liquefaction hazard and requirements for

remediation works under large earth

dams is potentially very significant.

9.0 ACKNOWLEDGEMENTS

The ERDC Centrifuge Research Center and the

Army Civil Works Earthquake Engineering

Research Program of the U.S. Army Corps of

Engineers sponsored this research. The

members of the ERDC centrifuge team who

helped conduct this work were Richard Burrows,

Lee Miller, Dr Gary Butler, Rodgers Coffing,

Wipawi Vanadit-Ellis, and Dr Mike Sharp. The

ERDC machinists contributed greatly to the

project. Drs William Marcuson and Michel

180
O'Connor were Directors of the Geotechnical

Laboratory. Permission to publish this paper by

the Chief of Engineers is gratefully

acknowledged.

10.0 REFERENCES

Arulanandan, A. and Scott, R.F. (1994) editors, Proc. Int. Conf. on Verification of Numerical

Procedures for the Analysis of Soil Liquefaction

Problems, at U.C. Davis; Volumes 1 & 2; Balkema.

Butler, G.D. (1999). A dynamic analysis of the

stored energy angular momentum actuator used

with the equivalent shear beam container, PhD

thesis, Cambridge University (in preparation).

Finn, W.D.L. and Yogendrakumar, M., Yoshida,

N. and Yoshida, H., (1986), "TARA-3: A

Program to Compute the Response of 2-D

Embankments and Soil-Structure Interaction

Systems to Seismic Loadings", Department of

Civil Engineering University of British

Columbia, Canada.

Finn, W.D.L., Ledbetter, R.H., and Marcuson,

W.F. Ill (1995). "Seismic Deformations in Embankments and Slopes", Symposium on Developments in Geotechnical Engineering - From Harvard to New Delhi, 1936-1994,

Bangkok, Thailand, A.A. Balkema, Rotterdam.

Finn, W.D.L. and Yogendrakumar, M., (1989)

"TARA-3FL; Program for Analysis of

Liquefaction Induced Flow Deformations",

Department of Civil Engineering, University of

British Columbia, Vancouver, Canada.

Ledbetter, R.H. (ed) (1991) Large centrifuge: a

critical Army capability for the future. Misc.

Paper GL-9 1-12, Dept. of the Army, Waterways

Experiment Station, Vicksburg, MS, May.

Ledbetter, R.H., and Finn, W.D.L. (1993).

"Development and Evaluation of Remediation

Strategies by Defocmation Analysis", ASCE

Specialty Conference on Geotechnical Practice

In Dam Rehabilitation, Raleigh, North Carolina,

April, pp 386 - 401.

Ledbetter, R.H., Steedman, R.S., Schofield,

A.N., Corte J.F., Perdriat J., Nicholas-Font J. and Voss H.M. (1994a) US Army's engineering

centrifuge: Design, Proc. Centrifuge 94, pp63-

68, Singapore, 3 1 Aug - 2 Sept, Balkema.

Ledbetter, R.H., Finn, W.D.L., Hynes, M.E.,

Nickell, J.S., Allen, M.G., and Stevens, M.G.

(1994b). "Seismic Safety Improvement of

Mormon Island Auxiliary Dam", 18th

International Congress on Large Dams, Durban,

South Africa.

Ledbetter, R.H., Steedman, R.S. and Butler,

G.D. (1999) Investigations on the behavior of

liquefying soils, Proc Int Workshop on Physics

and Mechancs of Soil Liquefaction, Baltimore

MD, Lade P.V. and Yamamuro J. A. (eds), 10-11

Sept 1998, Balkema, pp 295-306.

Peiris, L.M.N. (1999) Seismic modelling of

rockfill embankments on deep loose saturated

sand deposits, PhD Thesis, Cambridge

University.

Seed H.B. (1983) Earthquake-Resistant Design

of Earth Dams, Proceedings of Symposium on

Seismic Design of Embankments and Caverns,

May, ASCE pp 41-64.

Seed R.B. and Harder L.F. (1990) SPT-Based

analysis of cyclic pore pressure generation and

undrained residual strength, Proc. H Bolton Seed

Memorial Symposium, Vol. 2, pp 351 - 376,

BiTech Publishers Ltd, Vancouver.

Youd, L. and Idriss, I., Eds (1997) Workshop on

Evaluation of Liquefaction Resistance of Soils,

Proceedings, Sah Lake City, Technical Report

NCEER-97-0022, sponsored by FHWA, NSF

and WES, published by NCEER.

181
Figure 1. The WES centrifuge

VERTICAL SUPPORT WALLS

Figure 2. The ERDC Mk I earthquake actuator

182
Figure 3. Typical Equivalent Shear Beam specimen container

Figure 4. Shear sheets form the boundaries on the end wails of the ESB

183
79fi 1 V

water table at soil medium dense sand layer, T surface 1 RD = 75%

loose sand layer, RD = 50% i

140

160

15

796

water table at

soil surface medium dense sand layer, RD = 75%

loose sand layer, RD = 50%

365

160

15 Typical cross section through

Model series 1 and 2

Typical cross section through Model series 3

- 796 L

1 796 [

1

medium dense sand layer, RD = 75% water table at around 300 mm

depth

V

3

T

loose sand layer, RD = 50% 1

lead surcharge

15

water table at soil surface medium dense sand layer, RD = 75%

loose sand layer, RD = 50%

base layer, coarse sand

365

160

- 15 Typical cross section through Models 4a, 4b Typical cross section through Models 4c, 4d and 5

Figure 5. Cross sections through the different model configurations

184
surcharge comprising lead strips, totalling 586 lb (1 1 6 KPa)

V

phreatic u surface ?S

at soil surface ' ACC 7709

ACC 7771, PPT 15 eACC 12768, PPT31

dense sand layer, RX) = 76% ACC 12608, PPT8 e- eACC 12814, PPTIO

ACC 12801, PPTl 1 eACC 12815, PPT13

ACC 12804, PPT46 ACC 12806, PPT47

ACC 12817, PPT4I

loose sand layer, RD = 49% e- ACC I2S21, PPT50

PPT 38. ACC 12807 PPT 51, ACC 1281 1

slotted sheet and porous plates

Cross-section through Model 5b

80

160

70

j ^"-15

ACC 73 14 was mounted

horizontally on the bottom ring

Figure 6. Instrumentation for Model 5 b

15 Acceleration (%) 105

0

-50,6 0.65

-10

-15 ACC 7709 (near surface) and ACC 12814 (middle of dense layer) (fine) 15

10

5

0

-50J6 0.65

-10

-15 ACC 7709 (near surface) and ACC 12806 (top of loose layer) (fine) 1510

5

0

-15 ACC 7709 (near surface)

and ACC 12821 (middle of loose layer) (fine)

Time

(seconds)

Figure 7. Comparison of the measured surface acceleration with the acceleration at different depths,

Model 5b

185
ACC1 2804, top of upper layer

c

o

u-0

0)O

<

g

o

o<

g

TO0)

0)o

o<

g

ro

I0)0)o

o<

ACC12815, bottom of loose layer Time (seconds)

Figure 8. Comparison of measured accelerations at different depths, Model 2f

186
_ 150% , PPT 38, Upper layer

§ 100% i

If) <A<0o. 50% ^

0%

g 0 5 0.6 ' 0.7 0.8

i3 -50%

0.9 1.1 1.2 1.3 1.4 1.5 Time (seconds) _ 120% ^ 100%

i 80%

;

£ 60% .. £ 40%

Q. 20% + 0%

PPT 31, Top of loose layer

8

i5 -20%0i5 0.6 0.7 08 0.9 1

Time (seconds)

1.1 1,2 1.3 1.4 1.5

PPT 47, Middle of loose layer

Time (seconds)

-20%QL5 0.6 0.7 0.8 0.9 1.1 1.2 1.3 1.4 1.5

_ 120% — 100%

1 80%

2 60%

« 40%

Q. 20%

8 0%

PPT 46, Bottom of loose layer

Time (seconds)

uj -20% 015 0.6 0.7 0.8 0.9 1.1 1.2 1.3 1.4 1.5

20 Accelerometer 12815, Bottom of sand

15 -

Figure 9. Development of excess pore pressure in a sliailow bed, Model 2f

187
^ 40% , ^ 30%

3

j I 20% I

Q.

2^ 10%

^ 0%

o y

l5 -10% I _ 50% , - 40% i « 30% 1

(/) 0)5. 20%

PPT 50, Middle of loose layer

0.5

PPT 41, Middle of loose layer

1.5

QJO

X

LU

10% 1 0% LV 10% 9 i<rr '

.-iJJ^lLr^^,.,J,)|;l.J,''^''^^^^W^

0.5 1.5

60%

50%

Q)3 40%

0)

05 30% QJQ. 20%

ajo 10% -- Q.

(/) 0%

<n

0)O -10% 0

LU -20%

PPT 51, Bottom of loose layer

0.5 1.5 2 Time (seconds)

60%

50%

I 40%

« 30%

5. 20%

§ 10% " 0%

§ -10% (1^ ^ -20%

PPT 38, Bottom of loose layer

0.5 1.5 2

Time (seconds)

^5 Accelerometer 12811, Base of sand, smoothed

10

;

TO

O<

Time (seconds)

Figure 10. Development of excess pore pressure at over 500 KPa, Model 4d

188
„ 700

to Q.^ 500 -

5 300

in

0)^ 100

0) ^ -100 0

PPT 39, ' = 1037 KPa, 9.7 tsf

a. »1 iVf»1

.'"^

^-i'

" '.H'v

Time (seconds)

50

T

30

10 + S -10

0) T b -30- <

u -10 <

-20

0.2 0.4

ACC12806

0.6 0.8 1.2 1.4

0

-50 ^- Original record

20 ^ ACC12806

10 i 4

If':

i

i 1 RItered using Butterworth band pass, 250 and 10 Hz

I

Figure 1 1 . Development of excess pore pressure at initial vertical effective stresses of up to 1000 KPa

(10 tsf), Model 5d

Model 3c Base input motion (for reference)

Figure 12. Comparison between excess pore pressure development at different initial vertical effective

stresses

189
0 20

o

200

400

600

800

40

Ru (%)

60 80 100 120 140

Approximate

Field Depth (ft)

40

80

120

160

1000

1200

200

Figure 13. Comparison between maximum residual excess pore pressure and initial vertical effective

stress, all data

Approximate

Ru (%) Field

0 20 40 60 80 100 120 140 ^^P^^ (ft)

200 -

400 -

03 Q. 600

800

1000 -

1200

-J u

overconsolidated

OB

O

O

I normally

consolidated

40

80

120

160

200

Figure 14. Comparison between overconsolidated and normally consolidated data, all data

190
1 69m

139m

Zone in the foundation

where the research

findings would predict

improved performance

Zones in the foundation

(upstream and downstream)

where excess pore pressures may reach 100%, based on

current practice

Figure 15. Comparison of affected zones in the foundation

Figure 16. Deformations calculated using current analyses

191
Model

series Models in series Effective overburden

a, b, c, d, e, f

a, b, c, d, e

a, b, c, d

a, b, c, d

stress in loose

layer

1 tsf 2 tsf

3 - 5 tsf

7- 10 tsf

Depth of

prototype

(approx)

15 m

26 m

26 -40 m

54-63 m

Depth of

specimen

300 mm

525 mm

525 mm

525 mm

Notes

Nevada sand

Nevada sand

Nevada sand

with lowered

w.t. or surcharge

Nevada sand

with lead surcharge

Table 1. Summary of model tests

Model

Code

2a

2b

2c

2d

2e

2f

3a

3b

3c

3d

Overall

depth

(mm)

300

300

300

300

300

300

525

525

525

525

Relative

Density

44%

83%

50%

75%

49%

74%

50%

75%

49%

73%)

50%

75%

34%

73%

49%

77%

49%

79%

54%

80%

loose dense

loose dense

loose dense

loose dense

loose dense

loose dense

loose dense

loose dense

loose dense

loose dense

Ov' at mid- OCR Number

depth in of earth- loose layer quakes

(tsf)

1 1 3

1 1 2

1 1 5

1 1 4

1 2.5 4

1 2.5 4

2 1 2

2 1 3

2 1 3

2 2.5 4

Comments

Saturated to ground

surface. Saturated to ground

surface. Saturated to ground

surface. Saturated to ground

surface. Saturated to ground

surface. Saturated to ground

surface. Saturated to ground

surface. Saturated to ground

surface. Saturated to ground

surface. Saturated to ground

surface.

Table 2a. Detailed summary of experiments. Models 2a - 3d

192
4a 525 49% loose 3 1 4 Saturated to top of loose 80% dense layer only.

4b 525 56% loose 3 2.5 4 Saturated to top of loose 74% dense layer only.

4c 525 50% loose 4.7 1 4 Saturated to ground

75% dense surface. Lead surcharge.

4d 525 50% loose 4.7 2.5 4 Saturated to ground

68% dense surface. Lead surcharge.

5a 525 51% loose 7.4 1 4 Saturated to ground

72% dense surface. Lead surcharge.

5b 525 49% loose 7.4 2.5 4 Saturated to ground

76% dense surface. Lead surcharge.

5c 525 52% loose 9.2 1 3 Saturated to ground

75% dense surface. Lead surcharge.

5d 525 57% loose 9.2 1 1 Saturated to ground

80%) dense surface. Lead surcharge.

Table 2b. Detailed summary of experiments, Models 4a - 5d

Specific g

ravity 2.64

Maximum void ratio 0.757 (density 93.8 pcf)

Minimum void ratio 0.516 (density 108.7 pcf)

D50 0.18 mm (approx)

D,o 0.11 mm (approx)

Table 3. Nevada Sand (parameters as measured)

Density 1200 kgW

Viscosity 50 cs Specific Gravity 1.26

Composition 80% glycerine-water mix (by weight)

Table 4. Parameters for pore fluid (as measured)

193
I
SOIL AMPLIFICATION FACTOR FOR SEISMIC DESIGN OF BUILDINGS

By

Izuru 0KAWA'\ Masanori IIBA^\ Mitsumasa MIDORIKAWA^^

Shin K0YAMA'\ and Kenji MIURA^'

ABSTRACT

In this paper, a method for evaluating the

amplification property of earthquake motions due

to the surface soil. In the current Japanese building code, the site response effect is represented by three kinds of

soil classification, hard, intermediate and soft. This classification is simple enough to specify the

design seismic force for the soil conditions.

However, when soil investigation is conducted

and one may get numbers of information on soil properties and can evaluate some properties

consistent with the soil condition, it is extremely

rough to use only three categories. Therefore, we

bear it in mind to evaluate the necessary property

with given data specific to the site as much as

possible. This might be conformed to the

performance-based design method.

The basic concept of our proposal for seismic

design spectra for major earthquake motions is

1) Basic design spectra defined at the engineering

bedrock, and 2) evaluation of site response for geotechnical data of surface soil layers.

Meanwhile, it is not recommended for general

structures to be verified its seismic safety and/or

performance through the analyses using time

history of design groimd motions.

This paper proposes a method for evaluating

nonlinear amplification due to surface soil overlying engineering bedrock (EBR) using

design response spectrum. The proposed method

follows the procedures below.

(1) Represent the response characteristics of

surface soil on EBR by a multi-lumped mass model with shear springs and damping

coefficients.

(2) Calculate a period Ti, a mode shape {Ui} and

a modal damping factor hi for the first vibration mode.

(3) Evaluate the equivalent shear wave velocity

Vse of surface soil (4) Evaluate the amplification factor Ri of the

surface to EBR at Ti using an equivalent

surface single layer

(5) Evaluate the displacements in surface soil from Ri {Ui}, and hence evaluate the shear

strains. (6) With given nonlinear relations between shear

modulus (G), damping factors ( ^ ) and shear

strains, replace soil constants G and h by

those corresponding to strains. (7) Return to step (1) with revised G and ^

Iterate these procedures until Ti converges.

(8) After Ti is converged, calculate the response

distributions in surface soil. The results obtained by the proposed method are compared with those by "SHAKE" in terms

of the response spectrum at the surface as well as distributions of displacement, shear wave velocity and damping factor in the

surface soil. The applicability of the proposed

method for evaluating the nonlinear

amplification is confirmed through the

comparative studies.

KEY WORDS : Amplification factor, Seismic

design, Engineering bedrock, SHAKE

1)

2)

3)

4)

5)

Head, Construction Techniques Division,

Production Engineering Division, Building

Research Institute, Ministry of Construction

Head, Evaluation System Division, Codes

and Evaluation Research Center, BRI

Director, Intemational Institute of

Seismology and Earthquake Engineering,

BRI

Senior Engineering Staff, USEE, BRI

Director, Kobori Research Complex, Kajima

Corporation

195
1. INTRODUCTION

It is commonly adopted in seismic design of

structures to incorporate the dynamic property of

surface ground. The currently enforced Japanese

Building Standard Law specifies concept of soil classification to differentiate its effect to seismic

load. However, the classification is made so simply that the actual property of the surface soil may not be necessarily reflected with the

geotechnical data of the site, even if it is available.

It is important for engineers to design structures based on the field investigation result. Therefore,

the design code should facilitate such option to

utilize site specific property.

2. BASIC RESPONSE SPECTRUM AT

ENGINEERING BEDROCK

In our proposal, the earthquake load for seismic

evaluation is specified with earthquake ground

motion not with seismic force given in the current Building Standai d Law of Japan. The evaluation

of earthquake motion is represented with the

acceleration response spectrum in the following

formula.

1 1

S,{T) = ZG,{T)S,{T) (1)

Where, (T) : acceleration response spectrum

for evaluation, Z: seismic zoning factor, (T)

:

soil amplification factor, and 50(7") : basic

acceleration response spectrum at exposed

(outcrop) engineering bedrock.

The engineering bedrock is defined as a layer with more than approximately 400 m/s in shear wave

velocity. The basic response spectrum for major

earthquakes is illustrated in Figure 1

.

3. EVALUATION PROCEDURE OF

ACCELERATION RESPONSE SPECTRUM AT

GROUND SURFACE

We can compute the amplification factor (transfer

function) for one layer system supported by

flexible bedrock. Using the vertical shear wave

propagation and the hysteretic damping

assumption, the amplification factor at the first natural ft-equency is.

Similarly, for the second natural frequency &>2

,

H{CD2) = ^ 1 (3^/2)^ + ^ + a

where,

H{a)) : The amplification ratio between surface

and outcrop flexible base layer

a =

p : mass density of the layer

: mass density of the bedrock

: shear wave velocity of the layer

F^^ : shear wave velocity of the bedrock ^ : damping factor of the layer

The amplification to be determined here is the

ratio of response spectra. Therefore, the function

H(a)) is different fi-om the {&> = IttIT) .

It is easily supposed, however, that these two

values are related and take similar values. We

will not evaluate {T) theoretically but estimate

in an approximate manner employing the

expression of H{co) . The followings are the

method we discuss here. First of all, to evaluate

the acceleration response spectrum at the ground

surface, the amplification of surface soil deposits on the engineering bedrock is estimated. An

evaluation procedure by using the equivalent . linearization technique considering nonlinear soil properties is expressed in the followings. The

simplified analytical method in this procedure is proposed and examined in detail elsewhere [Miura

et al., 2000].

(1) Transformation ofresponse spectrum defined

at outcrop engineering bedrock: The earthquake

motion defined at the outcrop engineering bedrock

is given as an acceleration response spectrum with

5% damping ratio, 5^(7,^ = 0.05)

.

5*^(7,^ = 0.00) , A velocity response spectrum,

Sy (T,^= 0.00) , and a Fourier spectrum of

acceleration, Fa(T), have the approximate

relations as follows.

ITU

(2) Eigen value analysis ofsoil profile:

196
Subdividing the soil profile, a shear model of n

degrees of freedom is formed. The natural period,

7) , the vibration mode, t/, (normalized by the

value at the surface), and the modal damping

ratio, ^ , are obtained through the Eigen value

analysis.

(3) Equivalent shear wave velocity and

impedance: The surface soil layers are replaced to a uniform stratum with an equivalent shear wave

velocity, V^^ , and an equivalent mass density, , and an equivalent damping ratio, , which are calculated fi-om the properties of each soil layer.

^se^^jZVsA (3)

n. ,=1 Pe-^f.PA (4)

n ,=1 Where, V^. = ^/G. I /?, , H: the total thickness of the

surface soil layers, and, G, , /> , and di are shear

modulus, mass density, layer height, and damping

ratio at the i-th layer from the surface,

respectively. The impedance of a wave motion,

a , between the equivalent surface soil layer and

the engineering bedrock is expressed as follows.

pVre se PbKb

(5)

where, p^ : mass density at the engineering

bedrock, Vsi,: shear wave velocity at the

engineering bedrock.

(4) Amplification by surface soil: The

amplification of the uniform surface soil layer to the outcrop engineering bedrock is obtained by

using the one-dimensional wave propagation in frequency domain. The transfer function of the

surface soil layer and the engineering bedrock to the outcrop one are expressed as follows: a)

surface/outcrop engineering bedrock;

Gj(T,4g,a) , and b) engineering bedrock /outcrop

engineering bedrock; G^{T,^^,a). An equivalent

shear modulus, G„ , and an equivalent damping

ratio, , of each soil layer are calculated through

the G-y , ^-y relationships of soil properties

considering the nonlinear characteristics of the

surface soil layers.

(5) Acceleration response spectra at ground

surface and engineering bedrock: They are evaluated as follows.

5,(r,#=o)«F;(r)G,(r,#,,^)/(^) (6)

S,,{T,^=0)^F,{T)G,{T,4,,a)l{-^) (7) In-

(6) Modification ofacceleration response

spectrum at ground surface: To estimate the

acceleration response spectrum conservatively at the groimd surface, the spectrum is modified by

connecting the two peak points, by a straight line,

of the acceleration response spectrum

corresponding to the first and second modes of the

surface soil layer, in order to avoid excessive

reduction between these peaks.

The computations of amplification factors are summarized as follows.

Gs{T) = Gs,

o.sr. T < 0.87;

(8) = cT + d O.ST, <T< O.ST, = G5| 0.87; <r< 1.27;

1.27: <r<io

where.

c =

Gs^ - Gs^

0.8(7^-7^2^

d = Gs^-O.ScT^

(9) e = {G5^-1.0}/{l/1.2rj-0.1}

/ = G5, — 1 1.27J H. : thickness of the i-th layer

/?, : mass density of the i-th layer

Vs. : Shear wave velocity of the i-th layer

Go, : Shear modulus of the i-th layer ( G^, = p^Vj )

7; , 7^2 can be computed with the following

equations.

7; = (10)

T-'2 =

197
T^ : The first dominant period of the soil deposit

corresponding to the ground motion level.

T2 : The first dominant period of the soil deposit

corresponding to the ground motion level. G, : The shear modulus of the i-th layer as reduced

with the given table corresponding to the induced

shear strain.

Gs , Gs , and are given as follows.

Gs

1 -0.240^+1.270- + 0.03 J -0.040-' +0.61a

+ 1

1 ^-0. Bo' + 0.220 + 0.03 -0.02o^ +0.21o

1

- + 1

« 1-

1 -0.34o^+0.79o + 0.03j ' O.6I0

where,

a : Wave impedance ratio computed with the

following equation

a

where,

Pg : mass density of the engineering bedrock

Vg : shear wave velocity of the engineering

bedrock

h: damping factor computed with the following

equation

where,

^. : Damping factor of the i-th layer as modified

with the given table corresponding to the induced

shear strain. w. : Strain energy in the i-th layer computed with

the following equation.

' 2H,

'

Ui : the deformation of the i-th layer top to the

engineering bedrock

In some cases, the amplification factor Gs(T) at shorter period will be excessively decreased.

Therefore, for design, conservative values might

be assigned for Gs(T) , for example, greater than

1.0 or larger.

The deforamtion and its shape of the soil deposit can be roughly evaluated as follows.

4. EXAMPLES OF EVALUATION OF

ACCELERATION RESPONSE SPECTRUM AT

GROUND SURFACE

Figure 2 shows shear wave velocities, V^, in several soil deposits to be evaluated. The shear wave velocities are measured by the PS logging

method. The soil layer of of more than

approximately 400 m/s is selected as the

engineering bedrock. The nonlinear

characteristics of the surface soil layers are modeled according to the other publication

[Ohsaki et al., 1978]. The mass density of the

soils is around 1.6 to 2.0.

The amplification factors (transfer fimctions) of

the ground surface to the outcrop engineering

bedrock are shown in Figure 3. The figure

includes the results fi^om three analytical methods

in the following.

1) Transfer fimctions in case of multi-layers with

Vs by the linear analysis (indicated by "Linear"),

2) Ratios of acceleration response spectrum at the

ground surface to that at the outcrop engineering

bedrock, that are calculated by the computer

program SHAKE [Schnabel et al., 1972]

(indicated by "Equivalent"), and

3) Transfer functions obtained by the proposed

method (indicated by "Proposed").

The predominant periods of the surface soil layers

subjected to severe earthquake motions are longer

by 1.3 to 2.0 times than those to moderate

earthquake motions because of the nonlinear

behavior of soils. There are some differences

between the predominant periods of the proposed

method and Shake at Sites 3 and 4. The

amplification factors at Sites 3 and 4 by the

proposed method are a little less than those by

Shake. Figure 4 shows the acceleration response

spectra at the surface, which are obtained by the

proposed method and Shake. "Modified Co'Rr

198
also indicates the acceleration response spectrum

converted from the base shear coefficient of

buildings for medium soil deposits prescribed in the Building Standard Law of Japan. The

response spectra by the proposed method have

good agreement with those by Shake.

5. CONCLUSIONS

A response spectrum method is proposed for evaluating the nonlinear amplification of surface

strata overlying the engineering bedrock, and its applicability is examined. Concluding remarks are as follows.

(1) The proposed method does not require

acceleration waveforms, as do the rigorous

methods.

(2) The responses and the nonlinear characteristics

of the subsoil can be evaluated by the proposed

method when the design earthquake input motion

is prescribed only in the form of a response

spectrum instead of a time history. No significant

differences are observed between the results obtained by the proposed method and by the

rigorous method.

(3) The acceleration response spectrum on the

ground surface can be accurately evaluated by the

proposed method. However, when the soil constants change rapidly, such as at Site-4, the

acceleration response spectra obtained by the

proposed method are less than those obtained by

SHAKE in the range of the whole periods.

(4) The applicability of the proposed method is

confirmed from the viewpoint of practical seismic

design, if attention is paid to the peculiarity of the

proposed method described in the above term 3.

6. REFERENCES

(1) Hiraishi, H. and et al. (1999), "Seismic

evaluation of buildings by acceleration spectrum

at engineering bedrock (Part 1-Part 13)", Structures II, Summaries of Technical Papers of

Annual Meeting, AIJ (in Japanese).

(2) Hiraishi, H., Midorikawa, M., Teshigawara,

M., Gojo, W. and Okawa, I. (2000),

"Development of performance-based building

code of Japan -framework of seismic and

structural provisions-", Proceedings 12th World

Conference on Earthquake Engineering.

(3) Miura, K., Koyamada, K. and liba, M. (2000),

"Response spectrum method for evaluating

nonlinear amplification of surface strata", Proceedings 12th World Conference on

Earthquake Engineering.

(4) Ohsaki, Y., Hara, A. and Kiyota, Y. (1978),

"Stress-strain model of soils for seismic analysis",

Proceedings 5th Japanese Earthquake

Engineering Symposium, pp. 679-704 (in

Japanese).

(5) Schnabel, P. B., Lysmer, J. and Seed, H. B.

(1972), "SHAKE: A computer program for earthquake response analysis of horizontally

layered sites", Report No. UCB/EERC-72/12,

Univ. of Califomia, Berkeley, Calif

Figure 1 : Basic response spectrum at engineering bedrock

Figure 2: Shear wave velocity distribution at example sites Figure 3: Transfer fimctions of ground surface to engineering bedrock

Figure 4: Acceleration response spectra at groimd surface (h = 5%)

199
Figure 1 Basic response spectrum at engineering bedrock

Vs(m/s)

400 800

£-20

Q. Q-40

-60

SITE1

Period( linear) 0.46sec

Vs(m/s)

400 800

E -20

0) Q -40

-60

SITE2

Period{ linear) l.lSsec

Vs(m/s)

400 800

-60

SiTE3

Period(linear)

0.63sec

Vs(m/s)

400 800

Vs(m/s)

0 400 800

Vs(m/s)

400 800

I

-20

Q.

0) Q -40

-60

Engineering

Bedrock

SITE4

Period(linear)

0.17sec

Q.

0) Q -40

-60

PeriodClinear)

0.77sec

I

-20

£

Q.

CD Q-40

-60

\ SITE6

Period(linear) 0.31 sec

Figure 2 Shear wave velocity distribution at example sites

200
SITES

/

/

Period (s) Period (s)

Figure. 3 Transfer function of the ground surface above the engineerig bedrock

2000 2000 2000

"aisoo 1o,

til 000

o<

t 500 <

0

0.1

Surface (Equivalent) Surface (Proposed)

Engineering Bedrock

Modified Co- Rt (2nd soil type)

SITE4j

y

1 Period (s) 10

/ ^

SITE6|

pj' \\

7 ^

0.1 Period (s)

1 Period (s) 10

Figure 4 Acceleration response spectra at ground surface ( ^ = 0.05 )

201

CONSTRUCTION OF A BASE-ISOLATED HOUSE FOR OBSERVATION OF

ISOLATION EFFECTS DURING EARTHQUAKE AND WIND

by

Masanori IIBA", Mitsumasa MIDORIKAWA"', Yasuyuki YAMANOUCHP',

Masayoshi IKENAGA'' and Kenichi MACHIDA^'

ABSTRACT

Base isolation is very effective on keeping

safety during earthquakes. In case of houses,

it is difficult to elongate enough period of

buildings by laminated rubber bearing

isolators because of light weight. In addition to rubber bearings, sliding and

rolling devices are needed to be developed

or applied. Recently base-isolated buildings

have been constructed for considering

seismic safety. The number of base-isolated

houses is little. Because of cost for isolation

such as isolators, reinforcement of 1st floor

and foundation. In order to confirm the

effectiveness of base isolation during

earthquakes and influence on habitability

during winds, a base-isolated house has

constructed in Building Research Institute.

It is very important to accumulate the

observed date of the base-isolated house.

KEY WORDS: Base Isolation for Houses

Construction

Isolator Characteristics

Superstructure Characteristics

l.INTRODUCTION

Many structures suffered from the 1995

Hyogoken-Nanbu Earthquake. Some of

buildings have collapse at 1st or intermediate stories^'. Since the earthquake

occurred in the gray dawn (5:45am), there were many persons which were dead and

injured at their houses.

Two buildings with base isolation were

constructed in Kobe City. The observed

earthquake motion records in the buildings

showed the effect of base isolation on

reduction of acceleration response^'. After

the earthquake, many low- and medium-rise

buildings with isolators have been designed

and constructed. But the base isolation for

houses has not become popular^''^'.

Base isolation is very effective on keeping

safety during earthquake (response

reduction of acceleration and inter-story

drifts, and protection from overturning of

furniture, etc.). Laminated rubber bearings

have been used much for isolated buildings.

In case of houses, it is difficult to elongate

a period of buildings by laminated rubber

isolators because of the light weight of

houses. In addition to the rubber bearing

devices, sliding- and rolling-type devices

are needed to be developed or applied^'"^\

In order to accumulate the observed date of

1) Head, Structural Dynamics Division, Building

Research Institute, Ministry of Construction,

Tsukuba-shi, Ibaraki-ken, 305-0802, Japan

2) Director, International Institute of Seismology

& Earthquake Engineering, Building Research

Institute, Ministry of Construction, Tsukuba-shi,

Ibaraki-ken, 305-0802, Japan

3) Deputy Director General, Building Research

Institute, Ministry of Construction, Tsukuba-shi,

Ibaraki-ken, 305-0802, Japan

4) Oiles Corporation, Hakari-cho, Ashikaga-shi,

Tochigi-ken, 326-0327, Japan

5) Sumitomo Forestry Co., Ltd., Midorigahara,

Tsukuba-shi, Ibaraki-ken, 300-2646, Japan

203
the base-isolated house, a base-isolated

house has constructed in Building Research

Institute. Purposes to construction of the

3,185 I 3,185 I 3,640

10, 010 Plan

— (

+

2,730 3,640 3,640 ^60

10,010 2,410

Fig.l Plan of Base Isolated house

910

SI iding

System

Section

Fig.2 Section of Base Isolated House

base-isolated house are as follows;

a) to clarify issues in the design and

construction of base-isolated houses,

b) to observe effects of isolation on

response of houses during earthquake, and

c) to obtain behavior of the baseisolated house during strong wind.

2.0UTLINE OF BESE-ISOLATED

HOUSE

A plan of the base-isolated house, which has

been constructed at Building Research

Institute, in Tsukuba, is drawn in Fig.l.

A superstructure was built through a conventional construction method in Japan

(posts and beams of glued laminated timbers).

The outline of the house is as follows;

Building use Residence

Number of stories 2

Floor height(m) IF 2.925

2F 2.785

6.82

IF 69.56

2F 51.34

120.90

The house is a standard size in Japan, and its foundation is a mat one. To considering

stable bearing capacity to prevent from nonequal settlement for a long term, a soil improvement was made.

A section of the baseisolated house is shown

in Fig.2. Beams for

supporting a weight of

1" floor are made of

glued laminated timbers

with a large cross section of 210mm in width and 360mm in height. The floors,

which are made of

plywood with 35mm in

Eaves height(m)

Floor Area(m^)

Total Area(m^)

eaiii:210X350

\1 05X240

204
thickness are attached to beams to improve

the in-plane stiffness of floors. Isolators are

set between 1" floor and foundation (isolated

layer). The isolators with a friction system so

called the friction pendulum system (FPS) as

Upper ^herical Plate

MDvingEfevice \ Lower Sptencal Plate \ admg^I^^

^ Caverix^forDust

Fig. 3 Section of Isolator

= 3.2!cc I--!.7lor!f (sialic loiding)

Fie.4 Characteristics of Isolator

isolator

b a is e Fig. 5 Analytical Model

Table 1 Fundamental properties

direction X axis Y axis stiffness kgf/cm K1 9500 10000

K2 6700 5700

weight

WB 13547 13547

W1 16310 16310

W2 8679 8679

natural period 1st 0.361 0.362

2nd 0.166 0.175

shown in Fig. 3 are used in the house. The

isolator is consisted of double spherical

plates, and a sliding cylinder made of Teflon

material. The 14 isolators are installed. A

relationship between horizontal force and

displacement is drawn in Fig. 4. A coefficient

of friction is about 0.02, and a period from

length of the pendulum is 3.2 sec. The

allowable displacement is 340mm in one

side.

Fig. 5 presents an analytical model to simulate seismic response of the house. The

house is modeled to three degree of freedom

with shear deformation. The superstructure

and isolator are assumed to be linear and the

Acceleration(X-dir<;ction)l

EL CENTORO NS HACHINOHE NS JMA KOBE NS

100 200 300 400 500 600 ACCELERATIONCgal)

RF 1 ^ : 9-<?

Relative DisplacementC X-direction)

EL CENTORO : NS ] 2R _ I HACHINOHE

NS

•JMA KOBE NS 9~9

50 100 150 DlSPLACEMENT(mm)

Fig. 6 Acceleration and Displacement

ResDonse Durine Earthauake

205
NS Earthquake ground motions are 1940 El

Centre, 1968 Hachinohe Harbor NS and 1995

Japan Meteorological Agency at Kobe NS

records which amplitudes are adjusted to 50

cm/s. The maximum accelerations of 1" floor are about one third of the maximum ones of

input motions. The displacements of isolators are 100 to 160 mm and angles of each story

drift are less than 1/700.

3.CONSTRUCTION OF BASEISOLATED HOUSE

The construction period of the base-isolated

house is June to September in 1998. It took

around one week to setting isolators, beams

of 1" floor and coverage around isolated

Photo. 1 Soil Imorovement

Photo. 2 Isolators

layer. The period related to works for

isolation was not so much long. The

construction flow is shown in the following.

Subsoil exploration

1 Soil improving(Photo. 1)

i Bar arrangement

1 lsolator(Photo. 2)

I Beem Establishment(Photo. 3)

i Upper structure

i Entrance • approach

i Equipment piping

i G utter

i Complete view(Photo.4)

Photo. 3 Girders for 1st floor

Photo. 4 Overall view of Base

Isolated House

206
4.CHARACTERISTICS OF ISOLAOTR

In order to clarify characteristics of isolated

layer, a static loading test was carried out. Two static oil jacks were set between the 1"

Reactioft; pos

Section

Fig. 7 Setup for Loading Test

Fig. 8 Load vs. Displacement

(with Coverage of Rubber)

Fig. 9 Load vs. Displacement

(with Coverage of Rubber)

floor and a reaction wall, as illustrated in Fig.

7. Horizontal forces were exerted to the

isolated house and horizontal displacements were measured in each isolator. The

relationship between force and displacement

in X-direction is drawn in Fig. 8. In the

figure, the relationship with several

coefficients of friction. Much horizontal

force is need to starting slippage at the

beginning of stage because of static friction. In the region of small displacement, the

friction force is larger than that of coefficient

of 0.02. The reason is as follows. There is a rubber coverage around isolators protecting

from dust entering inside. The friction force

between the isolator and rubber coverage is negligible and the coefficient of friction

tends to be large. As the displacement will be

large, the influence of the friction force by

rubber becomes little. To investigate the

effect of the rubber coverage, the results

without it is shown in Fig. 9. Though there is the static friction at the beginning stage, the

static friction changes to the dynamical one

after starting sliding.

5.CHARACTERISTICS OF SUPERSTRUCTURE

To clarify characteristics of a superstructure

of the base isolated house, microtremor and

oscillator tests on the base isolated house were carried out. Since amplitudes of

external force on these tests are not so large,

the isolators does not move and the

superstructure is assumed to be fixed at the

base.

Resonant curves of the superstructure in the

X(NS) and Y(EW) directions by the

oscillator test are drawn in Fig. 10 and 11,

respectively. In the figures, amplitudes of

displacements at each level which are normalized by the force due to the oscillator,

207
and phase lags at each level to the oscillator

force are presented. The oscillator was set on

the center of the roof frame. The moving

weight is 36 kgf. The superstructure was

excited by sweep vibration, that is, variable

frequency with 3 to 12 Hz. The resonant

frequencies of the superstructure in X and Y

directions are 6.0 and 6.5 Hz, respectively.

Damping factors are 6.4 and 8.4%, which are calculated through the one by the square root method. From the oscillator test, the damping

factors are lager than those which are expected.

6.0BSERVATI0N SYSTEM FOR

ISOLATED HOUSE DURING

EARTHQUAKE AND WIND

To investigate the characteristics of the base

isolated house during earthquake and strong

winds, the observation system was installed.

A sensor list is presented in Table 2. Accelerometers are set on the each floor (to see in Photo. 5) and displacement transducers

are set several isolators (to see in Photo. 6).

Table 2 Sensor Lists for Observation

Channel no. TriKKer Position Sensor Direction Sensitivity

1 NS

2 1 St floor Acc. EW 2000gal/10V

34 NS

5 2nd floor Acc. EW 2000gal/10V

6 UD

7 Roof Acc. NS 2000gal/10V

8 EW

9 Igal Acc. 2000gal/10V

10 Base floor Vel. NS 200kine/10V

11 Vel. 5kine/10V

12 iKal Acc. 2000gal/10V

13 Base floor Vel. EW 200kine/10V

14 Vel. 5kine/10V

15 Base floor Acc. UD 2000gal/10V

16 Vel. 200kine/10V

17 NS

18 Base floor Oisp. EW 500mm/ lOV

19 NS

20 EW

21 Pole Wind direction 540° /V

22 9m/s Wind velocitv 60m/V

23 1 St floor Pressure NS Ig/cmVl V

24 2nd floor Pressure

losclllator Test (NS.600gal)| 1 ; — RF/FORCE(3-12Ki)

2F/FOFK:EC3-12m)

lF/FORCE(3-12Ht) — 8F/FORCe(3-l2Hz) fFragu«i<cr(Hz)

Fr«qu«ncy(Hz)

10 12

3 SO

315

270

225

ISO

JoscUlator Test(EW.600gal)|_

RF/FORCE(3-l2Hz)

2F/FORCE(3-l2Hi)

1F/FORCE(3-12Hi)

- BF/FOROE(3-l2Hi)

Fr.guancyCHz) FroQucncv(Hz)

Fig.lO Resonant Curve

(X-direction)

Fig. 11 Resonant Curve

(Y-direction)

208
Against the strong wind, a wind velocity and

direction will be measured in the system.

The earthquake observation wave records are drawn in Fig. 12. The earthquake with JMA

magnitude of 5.1 occurred at 26"' of March in 1999. The epicenter of the earthquake was

the north in Ibaraki-ken. The distance is about 100 km and the depth id about 50 km.

In Fig. 12, the acceleration records at each

floor, shear force calculated from the

accelerations and the displacement of the

isolator in the X-direction. The amplitude of

acceleration on the foundation and 1^' floor

are 41.7 and 46.8 gals, respectively. Though

the isolator shows the movement with the

Photo. 5 Accelerometer

Photo 6 Displacement Transducer

displacement of 1.22 mm, there is no remarkable effect of base isolation to the

response of the superstructure.

7.CONCLUSIONS

There are several issues to spread the

construction of base isolated houses. The

issues are to considering the effect of strong

winds to isolators and to decrease costs to construction of base isolated houses. In case

of light and/or low story houses, the dynamic

characteristics of houses against earthquakes

and strong winds are dependent on those of

isolators. To investigate the dynamic

behaviors of base isolated houses, it is necessary to accumulate the practical data

though the observation during earthquakes

and strong winds.

8.ACKNOWLEGDEMENT

The shaking table tests and date analysis were conducted under the committee on the

cooperative research "Development of

seismic isolated houses(chairman is Mr.

Shoichi Yamaguchi, president of TokyoKenchiku Structural Engineers). The

authors express our sincere thanks to the

members of the committee.

9.REFERENCES

1. Building Research Institute : A Survey

Report for Buildmg Damages due to the

1995. Hyogo-ken Nanbu Earthquake, 1996

March

2. The Building Center of Japan : Base

isolated Buildings -Technical development

and Earthquake Motion Observation

Records Part 2-, 1995 June (in Japanese)

3. Yamanouchi H., M. Midorikawa and M.

liba (Editor): Research and Development on

209
Base Isolation Technology for House

Survey of Knowledge and Shaking Table

Test on House Model, for Evaluation of

Performance on Base isolated House-,

Building Research Data, No. 89, 1997

March (in Japanese)

4. Hiroyuki Harada et al.: Studies on Response Control Technologies for Building

Structures Part 11., Summaries of Technical

Papers of Annual Meeting, 1996 Sep.,

pp. 735-736 (in Japanese), Research

committee on Development of Seismically

Isolated Houses: Research and Development

on Seismic Isolation Technology for Houses

-Design Guidelines for Seismic Isolation

Houses (Draft)- Building Research Data,

No. 94, 2000 March (in Japanese)

5. Building Center of Japan : Sheet of

outlines for base isolated buildings. The

Building Letter 1996 June, pp. 57-58 (in

Japanese)

6. liba, M., Yamanouchi, H., Midorikawa, M.,

Yamaguchi, S., Ohashi, Y.and Takayama,

M.: Research on Performance of Base

Isolated House, Proceedings of the Second

World Conference on Structural Control

(2WCSE), Kyoto, Japan, 1998, 2, pp.1 119-

1126

7. liba, M., Midorikawa, M., Kawai, M.,

Kawazoe, Y., Ikenaga, M. and Morikawa S.: Shaking Table Tests on Seismic Behavior of

Isolators for Houses, Proceedings of the lO"" Japan Earthquake Engineering Symposium,

Yokohama, Japan, 1988, pp. 2653-2658

8. liba, M., Midorikawa, M., Yamanouchi, H.,

Yamaguchi, S., Ohashi, Y. and Takayama,

M: Shaking Table Tests on Performance of

Isolators for Houses Subjected to Three

Dimensional Earthquake Motions,

Proceedings of the Twelfth World

Conference on Earthquake Engineering

(12WCEE), Auckland, New Zealand,

N0.1765/6/A, 2000 Jan.

210
100

50

0

-50

-100

2.52

1.5

1 0.50

-0.5

Acc:1 F-X(46.9gal

i

15 20 25 30 35 40

15 20 25 30 35 40

•Dis:1F-XE(1.23mm)

0 5 10 15 20 25 30

Fig. 12 Observed wave forms in 1999.3.26 earthquakes

35 40

211

Design Procedures for Structures with Supplemental Pampers

Fahim Sadek and Michael A. Riley"

ABSTRACT

Passive energy dissipation devices are used to reduce the damage from earthquakes by absorbing a portion of the earthquake-induced en- ergy in structures. Wide acceptance of these

devices in structures will depend on the availability of simplified methods for their analysis

and design. This study is concerned with: 1) investigating the effect of increased viscous

damping on the seismic response of structures; 2) assessing the accuracy of the linear static and

dynamic procedures recommended in the

NEHRP Guidelines for the Seismic Rehabilitation of Buildings (FEMA 273) for the design of

structures with velocity-dependent dampers; and

3) proposing modifications to the current procedures to improve their accuracy. Based on the

analysis of several single-degree-of-freedom

structures subjected to a large number of earth- quake records, the study identifies two limitations of the FEMA 273 procedures: 1) the use of

a constant reduction factor for the displacement

response of short-period structures; 2) the as- sumption of a harmonic response to compute the

peak velocity and the story and base shears. In most cases, these assumptions result in nonconservative estimates of the peak response and

design forces. Comparisons of the methods pro- posed in this study and in FEMA 273 indicate

that the former produces more accurate results. KEYWORDS: design codes; passive energy

dissipation devices; seismic codes; supplemental

damping; structural control; viscous dampers.

1 INTRODUCTION

Supplemental dampers, also known as passive

energy dissipation devices, can absorb a portion

of earthquake-induced energy in the structure and reduce the energy demand on the primary

structural members. The NEHRP Guidelines for the Seismic Rehabilitation of Buildings (FEMA

273) categorize these devices according to their

mechanical behavior as: 1 ) displacementdependent devices, which have forcedisplacement response characteristics that are primarily a function of the relative displacement;

2) velocity-dependent devices, which have

force-displacement response characteristics that are primarily a function of the relative velocity or the frequency of motion; and 3) other devices,

which cannot be classified as either displacement- or velocity-dependent.

FEMA 273 guidelines present simplified linear

static and dynamic procedures, as well as the more sophisticated nonlinear static and dynamic

procedures, for the analysis of rehabilitated structures incorporating these devices. This

study is concerned with investigating the influ- ence of velocity-dependent supplemental dampers on the seismic response of structures, evaluating the accuracy of the linear static and dynamic procedures presented in FEMA 273, and

proposing modifications to these procedures for the analysis of structures with velocitydependent dampers.

The paper includes a brief review of studies on

the influence of supplemental viscous damping

on the seismic response of structures, as well as the analysis procedures presented in FEMA 273.

Analyses of several single-degree-of-freedom

by

213

Research Associate, Southern Methodist University, Dallas, TX 75275; on Assignment, National Institute of Stan- dards and Technology, 100 Bureau Dr., Stop 86 11, Gaithersburg, MD 20899-861

1

^ Research Structural Engineer, National institute of Standards and Technology, 100 Bureau Dr., Stop 8611,

Gaithersburg, MD 20899-861

1
(SDOF) structures with different damping ratios under a large number of earthquake excitations were carried out. The results of the statistical analysis were used in developing linear static and dynamic procedures for the design of struc- tures with linearly elastic behavior and velocitydependent dampers. The method proposed in this study is compared with that recommended

in FEMA 273 and with time history analyses to

illustrate the accuracy of the proposed method.

2 PREVIOUS RESEARCH AND SEISMIC

PROVISIONS

The influence of increased viscous damping on

the seismic response of structures has been

studied by a number of investigators. Newmark

and Hall (1982) presented the effect of damping

ratio, P , in the range of 0.5 to 20 % on the median acceleration, velocity, and displacement

amplifications used to construct design spectra.

Their results indicated reduced amplifications

for increased damping. The effect of inherent

and supplemental damping on the earthquake

spectral displacement, SD , has been studied by

Ashour and Hanson (1987), who used damped

SDOF structures with natural periods, T, ranging

from 0.5 s to 3.0 s. The computed spectral dis- placement for each period was normalized to those for zero and 5 % damping ratios for each

record. Their statistical analysis led to the intro- duction of a reduction factor that reflects the

decaying pattern of the spectral displacement

with the increase in the damping ratio. Wu and

Hanson (1989) studied the elastic-plastic re- sponse of SDOF systems with large damping

and different ductilities. The study indicated

that the effect of damping on the inelastic re- sponse is similar to its effect on the elastic re- sponse.

The above studies indicate similar conclusions

regarding the reductions in earthquake spectral

displacements with an increase in viscous

damping. No consideration, however, was given

to the influence of increased damping on the

absolute acceleration response which, as dis- cussed later, is the key parameter for computing

the seismic forces and base shears in structures with passive energy dissipation devices.

The NEHRP Guidelines for the Seismic Rehabilitation of Buildings (FEMA 273) present a method for constructing design spectra for a given damping from the 5 % damped spectra.

The procedure is to divide the spectral ordinates

in the constant acceleration and constant velocity regions by factors, and Bj, respectively,

corresponding to the specified damping ratio. These factors are based on the recommendations

of Newmark and Hall (1982). FEMA 273 also

presents the first comprehensive method of

analysis and design for structures with passive

energy dissipation devices. The guidelines rec- ommend the use of a simplified linear static pro- cedure (LSP) or a linear dynamic procedure

(LDP) for structures with linear behavior and

with displacement- and velocity-dependent

dampers. In addition, for structures with nonlinear behavior, the guidelines recommend a nonlinear static or a nonlinear dynamic procedure.

The main assumptions of the LSP and LDP can

be summarized as follows: 1) use factors B^ or Bj to reduce the base shear (if LSP is used) or

spectral acceleration (if LDP is used) for 5 %

damping; 2) assuming a harmonic response, compute the velocity between the damper ends

by multiplying the displacements by the fundamental frequency (LSP) or the modal displacements by the modal frequencies (LDP); and 3)

compute design shears as the maximum of the

following stages of deformation: at maximum

drift, at maximum velocity and zero drift where

only the viscous forces in the dampers are applied to the structure, and at maximum floor ac- celeration where the shear forces are computed

as the sum of the forces at maximum drift multiplied by CFj and the forces at maximum velocity multiplied by CF2 . The expressions for CFj and CF2 as a function of the effective

damping ratio, Peff ^ ^""^ given assuming a harmonic motion as:

CF,=cos[tan-'(2p,ff)]

CF2^sin[tau~^(2p,ff)]

214
3 ANALYSIS

The reduction factors presented by Newmark

and Hall (1982), Ashour and Hanson ( 1987), and

Wu and Hanson (1989) can be used to modify

the displacement or drifts of structures with sup- plemental dampers. They cannot, however, be

used to compute the design base shears. The

reason is explained in the following:

For a conventional structure modeled as a SDOF

with mass m , damping c, and stiffness k subjected to an earthquake excitation, the equation

of motion is given by:

mx,(t) + cx(t) + kx(t) = 0 (2)

where x,(t) is the absolute acceleration which

is equal to x(t) + Xg(t) ; x(t}, x(i), and x(t)

are the relative displacement, velocity, and acceleration, respectively; and ig(l) >s the

ground acceleration. For design purposes, the

base shear, V, is typically computed as the peak

force in the spring, which is equal to:

V^k\x(t)\ =ma)^ SD = m PSA (3)

where a) is the natural frequency and PSA is the

pseudo-acceleration ^co^ SD . When the SDOF

structure is equipped with a supplemental

damper, the base shear is computed from the sum of the forces in the spring and damper.

Therefore,

V = \kx(t) + cift)l^ ^\mx,(t)l^=mSA (4)

Here c represents the sum of the inherent and

supplemental damping coefficients. For this case, the base shear is equal to the mass times

the peak absolute acceleration, SA, rather than

the pseudo-acceleration PSA. Note that for zero damping, PSA is equal to SA and for small

damping ratios (up to 10 %), the two are ap- proximately equal and may be used inter- changeably. For larger damping ratios, however, the difference between PSA and SA is significant and the pseudo-acceleration cannot re- place the absolute acceleration for computing

base shears.

To investigate the effect of the damping ratio on

the relative displacement and absolute acceleration response of structures, linear SDOF systems

with periods ranging from 0.1 s to 4.0 s with

increments of 0.1 s and damping ratios of 2, 5, 10, 15, 20, 30, 40, 50, and 60 % were considered

in this study. The structures were subjected to a

set of 72 horizontal components of accelerograms from 36 stations in the western Unites

States. Refer to Sadek et al. (1999) for a complete list of these records. For each damping,

the relative displacement and absolute acceleration response ratios are computed as the ratio of

the peak responses of the structure to their counterparts with a damping ratio of 5 %. The mean displacement and acceleration ratios for the 72 records are presented in Figures 1 and 2, respectively. The figures show that increasing

the supplemental damping results in a further reduction in the displaceinent response. The

effect of damping is more pronounced in the

velocity region (structures with mid-range periods). While larger damping ratios (greater than

approximately 40 %) provide further reductions

in the displacement response, the additional re- ductions are not that significant and the in- creased damping adversely affects the absolute

acceleration response and consequently, the

shear forces, for long-period structures. For

short-period structures (T < 1 s), however, the

figure indicates that increased damping results in smaller absolute accelerations and consequently

smaller shear forces.

Since the force in a viscous damper depends on

the relative velocity, it is important to study the

effect of increased damping on the relative velocity response of structures. Therefore, for each record, the spectral velocity, SV , was

computed for different periods and damping ra- tios and divided by the pseudo-velocity,

PSV - coSD . From the mean values of

SV / PSV for the 72 accelerograms. Figure 3, it may be concluded that assuming the spectral

velocity to be equal to the pseudo-velocity (harmonic response) is valid only for periods in the

neighborhood of 0.5 s. For periods shorter than

0.5 s, the peak velocity is smaller than the

pseudo-velocity while for periods longer than

0.5 s, the peak velocity is larger and increases as the period and damping ratio increase. A regres215
sion analysis was used to establish the following

relationship between the spectral velocity and

pseudo-velocity responses:

SV = a^ PSV = a„ co SD

where

a„ =1.095 + 0.647 p-0. 382

b„ =0.193 + 0.838p -0.621P'

(5)

(6)

Figure 4 presents the mean ratio of the spectral

to pseudo-acceleration response {SA/PSA). The

figure confirms the discrepancy between SA and

PSA especially for structures with long-periods

and large damping. The relationship between

SA and PSA can be obtained using regression

analysis as:

SA = (l + a^ T^^ )PSA

. ... .}1.895 a, = 2.436 P'

=0.628 + 0.205

p

4 PROPOSED PROCEDURES

(7)

Based on the analysis presented in the previous

section, improved linear static and dynamic pro- cedures for analysis of structures with velocitydependent dampers were developed. In the pro- posed procedures, three damping factors are in- troduced: a displacement damping factor, , that can be used to reduce displacements and

drifts; a velocity damping factor, , that can be

used to compute relative velocities; and a force

damping factor, , that can be used to amplify

or reduce story and base shears. The factor , which is based on Figure 1, is presented in tabular form in Sadek et al. (1999). a^, is given

by Equation (6) while can be derived as follows assuming that SA = PSA for p =5 %:

Gn = SA(P) _ SA(p) PSA(P)

SA(5%) PSA(p) SA(5%)

A , co^SD(P) , „ (8)

d + aj"^)

co^SD(5%)

= (l + aj"^

The three damping factors may be used in design as follows: for the structure without sup- plemental dampers, compute the base shear (if

using LSP) or the spectral acceleration (if using

LDP) from the 5 % damped design spectrum,

and compute the floor displacements, inter-story

drifts, and story shears as usual. Multiply the

displacements and drifts by the displacement

factor to determine the actual deformations,

multiply the story shears by the force factor

to compute the actual shear forces, and compute

the damper forces as the damping coefficient

times the relative velocities, which are computed

as the product of the relative displacement, natural frequency, and velocity factor . Although

simple, this method does not explicitly include

the contribution of the damper forces in the

computation of design shears. Therefore, erro- neous results may be obtained for multi-story

frames with a non-uniform distribution of dampers. Consequently, another method is presented

in this study. This method is similar to the one

in FEMA 273 which includes the influence of

the damper forces on the computed design

shears directly; thereby, accounting for the dis- tribution of the dampers along the height of the

building.

The method assumes that the velocitydisplacement response in the quadrant with the

peak displacement, SD, and peak velocity, SV,

can be approximated by an ellipse. The accuracy of this assumption has been verified using

four structures subjected to two earthquake rec- ords (see Sadek et al., 1999). The velocitydisplacement relationship, thus, can be ex- pressed as:

x(t)

SD

x(t)

SV

= 1 (9)

Maximizing the design force

F(t) = kx( t ) + cx(t ) subject to the constraint of

Equation (9) will result in the peak forces corre- sponding to the maximum acceleration given as: V = Ci kSD + C2cSV =Ci + C2 (10)

where and /\, are the shear forces corre- sponding to the maximum drift and velocity,

respectively, and C/ and C2 are given by:

216
The accuracy of the proposed method of computing the design shears has been verified using

a comparison with time history analysis. Refer

to Sadek et ai. ( ! 999) for details.

As Equation (10) indicates, the design shears

will include the contributions of the peak re- storing forces and peak damping forces. Figure

5 shows the percent contribution of each to the

total design forces for damping ratios of 10, 20,

30, and 40 %. The figure indicates that for damping ratios up to 10 %, at least 90 % of the

design forces are due to restoring forces, while

for larger damping ratios, the contribution of

damping forces is more significant. This is es- pecially true for long-period structures, due to the effect of increased velocities as indicated by

Figure 3.

Based on the above analyses and discussions,

the proposed method for the linear static (LSP)

and linear dynamic (LDP) procedures can be

summarized in the following steps:

1. Compute the fundamental frequency using

any of the methods presented in FEMA 273

(LSP) or perform a modal analysis to determine the natural frequencies and mode

shapes of the structure (LDP). Assume an

effective damping ratio, ^eff ^ ^^^^ fundamental mode (LSP) or in each mode (LDP)

and determine the displacement, velocity,

and force damping factors (a^ , , and

, respectively) using table (Sadek et

al., 1999) and Equations 6, 7, and 8.

2. Compute the base shear as the product of the

5 % damped base shear and and distribute the forces, , along the height of the

structure using the method presented in FEMA 273 for conventional structures (LSP) or multiply the 5 % damped spectral

acceleration by for each mode (LDP),

and compute the design displacements and

inter-story drifts.

3. Calculate the effective damping ratio,

/^eff ^ using the method outlined in FEMA 273 and

iterate on steps 1 through 3 until the desired

accuracy is achieved.

4. Compute the relative velocity between the damper ends as the product of inter-story

drift, fundamental frequency, and a^, (for

each mode if using LDP). Calculate the damper forces as the product of the damping

coefficient and the relative velocity. Esti- mate the forces acting on the structure corre- sponding to the maximum velocity, , due

to the damper forces.

5. Compute the design shears corresponding to the maximum floor accelerations from

Equations (10) and (11). Coefficients Cy

and Ci should be computed for each story

(and for each mode if using LDP) rather than for the complete structure. The final design shears will be selected as the larger

of the forces computed at maximum acceleration or at maximum drift (step 2) multiplied by a^/a^ or 1 + a^T^'^ . This in- sures that the computed design shears are not less than those for 5 % damping multiplied by , as recommended in the previ- ous section.

The iterative procedure suggested in step 3 is used since the effective damping ratio, Peff^

depends on the displacement in the structures as recommended by FEMA 273. It should be noted

that FEMA 273 also presents an expression for estimating P^jj- (see Section 9.3.4.1 of FEMA

273). If using LDP, the modal displacements,

velocities, and story shears should be combined

using an appropriate method such as the square

root of the sum of squares (SRSS) or the complete quadrature combination (CQC).

5 COMPARISONS

To verify the analysis procedure proposed in this study, it was compared with that presented in FEMA 273, using several SDOF structures. Both procedures are compared with the average

217
results of time history analyses using the 72 ac- celerograms to assess their accuracy. For a SDOF structure with period T, effective weight

W, and no supplemental damping, the design

base shear is given as (FEMA 273):

^uncontrolled = C1C2C3SJV = aW (12)

where a = CjC jC^S^ : is the spectral acceleration at the fundamental period; and C/, C2

,

and are modification factors that can be

found in Section 3.3.1.3 of FEMA 273. When

the structure is equipped with supplemental

dampers, the three key design parameters are the

displacement, velocity, and base shear. For the

derivations of those three parameters using the

procedures presented in FEMA 273 and in this study, refer to Sadek et al. ( 1 999).

Figures 6-8 present comparisons between the

methods proposed herein and in FEMA 273 for the design displacement, velocity, and base

shear, respectively, for effective damping ratios of 10, 20, and 30 %. The figures also show the

results from the time-history analyses. For the

FEMA 273 procedure, the plots are presented

for ground motion at rock or hard rock sites with

spectral response acceleration of l.Og at short

periods and 0.4g at one-second period. The

three figures show the difference between the

two procedures (FEMA 273 and this study), es- pecially for larger damping ratios, and indicate

that the proposed method tends to be more accurate than that in FEMA 273.

Figure 6 shows that the two methods give dis- placements that are close to each other for long

periods. For shorter periods, however, the

method presented in FEMA 273 significantly

underestimates the displacements. This study

shows that for periods less than 0.5 s, the reduction factors vary substantially and are larger for shorter periods. Figure 7 shows that the two

methods yield velocities close to each other for periods less than 1 .0 s. For periods longer than

1.0 s, the FEMA 273 method gives velocities

and consequently, damper forces that are much

smaller than those predicted by a time history

analysis. Figure 8 shows that the method pre- sented in FEMA 273 does not reflect the increase in design base shears for structures with

long-periods and large damping ratios.

6 CONCLUSIONS

The overall objectives of this study were: I) to investigate the effect of increased viscous

damping on the seismic response of structures,

2) to assess the accuracy of the linear static (LSP) and linear dynamic (LDP) procedures

recommended by FEMA 273 for designing

structures with velocity-dependent passive en- ergy dissipation devices, and 3) to propose

modifications to the current design procedures to achieve better accuracy and reliability. The

findings of the study can be summarized as fol- lows:

• Increasing damping in structures allows more input seismic energy to be dissipated,

which generally reduces the response of

structures. The reduction, however, depends

on the structural period and the amount of

supplemental damping. For example, the ef- fect of damping on the displacement re- sponse is more pronounced in the velocity

region (structures with periods in the range

of 0.3 s to 0.5 s). For long-period structures, an increase in damping further decreases the

displacement response, but increases the absolute acceleration response and consequently seismic forces.

• Design forces should include the contributions of the peak restoring forces and peak

damping forces. The study indicates that for small damping ratios, approximately 90 %

of the design forces are due to the restoring

forces, while for larger damping ratios, the

contribution of damping forces is more significant. This is especially true for longperiod structures, due to the effect of in- creased velocities.

• The LSP and LDP recommended by FEMA

273 have the following limitations: 1) they use a constant reduction factor for the dis- placement response in the acceleration re- gion of the spectrum. This assumption can

result in a non-conservative design in short- period structures, 2) they assume a harmonic

response to compute the peak velocity, i.e.,

218
the peak velocity is equal to the pseudovelocity. This assumption results in accurate velocities, and consequently damper

forces, only for structures with periods close

to 0.5 s. This study, however, shows that for structures with other periods, the pseudovelocity should be multiplied by a factor that depends on the structural period and the

damping ratio, and 3) they assume that the

structure undergoes a harmonic motion with an amplitude equal to the peak displacement

and a frequency equal to that of the fundamental (or a given) mode. Consequently,

the same coefficients are used in each story

to compute the design forces. This study

shows that assuming an elliptical peak dis- placement-velocity response and using dif- ferent coefficients for each story to compute

the design forces is more accurate.

Based on the analysis of SDOF structures subjected to 72 earthquake records, modifications to the LSP and LDP presented in FEMA 273 are recommended. Comparisons between the methods proposed in this study and in FEMA 273

show that the method presented herein is more

accurate when compared with the time-history

analysis of several structures.

7 REFERENCES

Ashour, S. A., and'Hanson, R. D. (1987). "Elastic seismic response of buildings with supplemental damping." Report JJMCE 87-1, Dept. of

Civil Engrg., The University of Michigan, Ann

Arbor, Michigan.

FEMA 273 (1997). "NEHRP guidelines for the

seismic rehabilitation of buildings.'" Building

Seismic Safety Council, Washington, D.C.

Newmark, N. M., and Hall, W. J. (1982).

"Earthquake spectra and design." Monograph

Series, Earthquake Engrg. Res. Inst., Oakland,

California.

Sadek, F., Mohraz, B., and Riley, M. A. (1999).

"Linear static and dynamic procedures for structures with velocity-dependent supplemental

dampers,'" Technical Report NISTIR 6329, National Institute of Standards and Technology,

Gaithersburg, MD.

Wu, J., and Hanson, R. D. (1989). "Study of

inelastic spectra with high damping.'" J. Struct. Engrg., ASCE, Vol. 1 15(6), pp. 1412-1431.

<0

0) a:

0)E

O)

Q. Q

1.4

12

1 --

0 8 --

0.6 --

0 4

P=0.02 P=0.05

P=0.tO •=-p=0.15

P=0.20 -6..p=o.30 P=0.40 ••i- p=0.50 P=0.60

a a G a e e i

02 I. . . . I

. ' . I I I -+-

1.6

0 0.5 1 1 5 2 2.5 3 3 5 4 Period (s)

Figure 1. Mean Displacement Response Ratios

03

OH

14 _-

1.2 -

0,8

0.6

0 4

P=0.02

p=0.10 -o- —^P=0.20 - e- ^^p=0.40 - ^-^p=0.60

P=0.05

P=0.15 P=0.30

p=0.50 A

4

/

/^.^^"^

^ ^

a' a'

a

l\ rj a

'° ' "

1

' ' '

' 0 0,5 1 1.5 2 2 5 3 3.5 4 Period (s) Figure 2. Mean Acceleration Response Ratios

219
(3=060

0 0,5 1 1.5 2 2.5 3 3 5 4 Period (s)

Figure 3. Mean Spectral to Pseudo

Velocity Ratios

3

Q 2.5

1 p=0.02 P=0.05

; —^p=O.I0 • =- P=0.1.'i

--

1 -«-p=0.20 o- P=0.30

" -^p=0.40 • ' p=o..so

' " p=0.60

/

^

/^j: a a a"

, o -o o

- ' ' '

1

' ' '

1

' 1' '

1

'

' 1 ' 1 ' 1 ' 0 0 5 1 1.5 2 2.5 3 3.5 4 Period (s) Figure 4. Mean Spectral to Pseudo

Acceleration Ratios

Period (s) Period (s) Figures. Contribution of Restoring and Figure 6. Design Displacement for

Damping Forces to Total Design Forces SDOF Structures

220
221

SEISMIC DESIGN GROUND MOTIONS FOR STRAIT-CROSSING PROJECTS IN JAPAN

Masahiko Yasuda'', Keiichi Tanuiia'

ABSTRACT

In order to establisli rational seismic design for

strait-crossing projects in Japan, it is most

essential to appropriately consider the efTects of

near field ground motions of large earthquakes.

For tliis purpose, we study the applicability of

semi-empirical and stochastic Green's function

techniques to estimate near field ground motions,

by synthesizing ground motion for the

large-scale scenario earthquakes, in which

earthquake source parameters are systematically

changed. Based on the results of analysis,

variation of ground motion resulting from

uncertainty of source parameters is discussed.

Key words: strait-crossing project, seismic

design, ground motion, Green's

function, earthquake source parameter

INTRODUCTION

Feasibility of super long-span bridges crossing

straits is being studied in Japan to meet the goal

of constructing a multi-nuclei nation as set forth

in the Comprehensive National Development

Plan (National Land Agency, 1998). Some of

these bridges will be located over or nearby the

fault planes of past large earthquakes. For

instance, the 1923 Kanto Earthquake (M/=7.9,

Mj is the Japan Meteorological Agency

Magnitude) and 1944 Tonankai Earthquake

{Mj=l .9) occurred in Tokyo Bay and Ise Bay

regions, respectively, and two highway routes

are being planned to cross the mouths of these

two bays.

by

', Shojiro Kataoka^' and Yoshihiro Nakao^'

For establishing seismic design ground motions

of large structures, seismic risk analysis and

attenuation relation of ground motion have been

widely used in Japan including the Trans-Tokyo

Bay Highway Bridge (Public Works Research

Institute. 1991) and Akashi Kaikyo Bridge

(Honshu-Shikoku Bridge Authority, 1988). In the seismic risk analysis, an attenuation equation

is usually incorporated to estimate ground

motion intensity from an expected earthquake

magnitude and distance. Although an

attenuation equation is a simple and effective

tool of engineering practice to predict ground

motion, it should be noted that the applicability

of this technique to the near field of large

earthquake is limited, because it is an empirical

formula and does not usually contain strong

motion data obtained in the near fields of large

earthquakes. Appling attenuation relation to the near field by substituting small value for distance may yield excessive ground motion for seismic design of structures.

The semi-empirical ground motion synthesizing

method (e.g., Hartzell, 1978; Irikura, 1983), in which ground motion record from a small event

is used as Green's function, has recently been

applied to earthquake engineering. This

technique can incorporate complicated

earthquake source mechanism into calculation,

1 ) Formerly, Director, Earthquake Disaster Prevention Research Center, Public Works

Research Institute, Ministry of Construction,

Tsukuba-shi, Ibaraki-ken 305-0804 Japan

2) Head, Ground Vibration Division, ditto 3) Research Engineer, Ground Vibration Division,

ditto

223
and has an advantage of estimating near field motions. On the other hand, this method requires

various source parameters, and most of them are difficuh to be determined accurately, especially

for a future earthquake.

In addition to those facts, the semi-empirical

method requires an actual small event record,

however, it is obvious that an appropriate record

is not always available. To make up this disadvantage, stochastic synthetic method has

been proposed (e.g., Kamae, et al., 1991), in which stochastically simulated small event motion is used as a Green's function.

To establish rational seismic design ground

motions for strait-crossing projects, we are studying the applicability of semi-empirical and

stochastic Green's function techniques at the

Public Works Research histitute. hi this paper

we simulate ground motion at Tokyo Bay region,

assuming the Kanto Earthquake, which will be

most influential over this area. We employ

geometric parameters of the 1923 Kanto

Earthquake and change dynamic source parameters variously, and examine influence of

uncertaint)' of source parameters on the

synthesized ground motion, as a basis of

developing design ground motions for strait-crossing projects.

GROUND MOTION SIMULATION USING

SEMI-EMPIRICAL GREEN'S FUNCTION

TECHNIQUE

To simulate ground motions from a large event,

ground motion record from a small event such

as foreshock or aftershock that occurred within

the fault area of large event is used as Green's

function. The fault plane of the target event is divided into sub-faults so that their sizes

coincide with the fault area of small event. The

Green's functions are superimposed considering

time delay due to fault rupture from the

hypoccnter to a sub-fault and wave traveling

from each sub-fault to the estimation point. An

advantage of using small event record as Green's

function exists in that complicated path and site effects are incorporated automatically into ground motion estimation, which has large

influence over the accuracy of simulating short

period ground motions.

Temporal and spatial variations of dislocation

are considered here by introducing T„,„{(o) and

CO,,,,,, respectively, and the Fourier transform of

displacement from a large event u(co) is expressed as:

A' jV ''(^^)=X X ^""(^)^ U^{co)QXl(~iCOT )

(1)

where,

r,„„ : Function representing the difference

in source time function between

small and large events (Irikura,

1986)

. COT

sin —

r„H=i + (A^-i)-

COT2

sin- on

COT

exp - / COT COX 2 2{N-\)N,]

(2)

(Onm ' Weight function representing spatial

variation of dislocation

Wo : Frequency component of ground

motion from small event

T,„„ : Time delay due to fault rupture from

hypocentcr to sub-fault and wave 224
traveling from sub-fault to estimation point.

A': Number of sub-faults t: Rise time

A'l : Coefficient to avoid the

synthetically generated periodicity

The source time function after Irikura (1986) is used in the present study, which is shown in

F-igure I

.

VARIATION OF EARTHQUAKE

SOURCE PARAMETERS

Target Event and Basic Source Parameters

We employ a source model of the 1923 Kanto

Earthquake proposed by Wald and Somerville

(1995). The major parameters of this model are

listed in Table I. Ground motion at Kannonzaki

is synthesized in this study. We use a strong

motion data from the 1987 Chiba-ken Toho-oki

Earthquake (Mj=6J) as a small event record,

which was obtained on the siltstone at Kannonzaki. The fault plane of 1923 Kanto

Earthquake, and locations of epicenters of the

Kanto and Chiba-ken Toho-oki Earthquakes and

Kannonzaki are presented in Figure 2. Figures 3 and 4 show acceleration waveform and

acceleration response spectrum (damping ratio h=0.05) of Kannonzaki record, respectively.

Since the seismic moment of the Chiba-ken

Toho-oki Earthquake is calculated as 7.05x10^^

[dyne cm] after Harvard University (1999), the

fault plane of Kanto Earthquake is divided into

5x5 sub-faults in the following analysis,

according to the seismic moment ratio between

two earthquakes.

Source Parameters Changed

We systematically change fault rupture mode

and location of asperity to examine their influence on the synthesized ground motion

(Tamura et al., 1999). We assume lateral rupture

and radial rupture as illustrated in Figure 5, i.e.,

5 cases for lateral rupture and 15 cases for radial rupture. According to Somerville et al. (1999),

the area of asperity that has twice as large as average slip is 22% of total fault plane area, and

we assign asperity for 6 sub-faults out of 25

sub-faults. Note that their result was derived

from analysis of 15 crustal earthquakes, and

further study may be necessary to apply it directly to the inter-plate Kanto Earthquake. We

employ 2x3 and 3x2 asperities as shown in Figure 6. For each asperity shape, we assume 9

patterns of asperity distribution. Consequently,

we set up (5+1 5)x(9+9)=360 cases for numerical simulation at this stage.

In addition to the above-mentioned asperity

models, we also employ the variable-slip rupture

model of the 1923 Kanto Earthquake proposed

by Wald and Somerville (1995) for reference.

Note that the slip vector is determined for each

sub-fault in this model, however we use the slip

amplitude and ignore slip angle in the present

study. When we use this model, we assume

bilateral rupture from the hypocenter that is indicated by a star in Figure 7, and other source parameters are set as the same as shown in Table

1.

Since the ground motion estimation point

(Kannonzaki) is located over almost the center of fault plane as shown in Figure 2, some of

assumed simulation cases yield similar results when symmetric rupture mode and location of

asperity is assigned. Besides this, in the previous

study (Tamura et al., 1999), we confirmed that

the horizontal location of 2x3 asperity and

vertical location of 3x2 asperity do not affect much calculation result, respectively.

Considering those facts, we reassign rupture

225
mode and asperity locations as indicated in Figures 8 and 9. Tlien, numerical simulation

cases are reduced to 1 24 cases (Table 2).

VARIATION OF GROUND MOTION DUE

TO UNCERTAINTY OF SOURCE

PARAMETERS

Acceleration response spectra of the 124 cases

are overlaid in Figure 10, where damping ratio is taken as /7=0.05. Variation of spectral amplitude

depends on natural period, and it ranges from 3

to 6 times. We should be attentive that this variation is inevitable when we infer ground

motion for a future earthquake that has uncertain source parameters. Figure 1 1 shows acceleration

response spectrum using the variable-slip

rupture model that is shown in Figure 7. Comparing those two figures, result from the

variable-slip rupture model gives almost upper

bound of variation of synthesized ground

motion.

GROUND MOTION SIMULATION USING

STOCHASTIC GREEN'S FUNCTION

TECHNIQUE

Procedure of Ground Motion Simulation

Here, the ground motion due to each sub-fault is simulated stochastically and employed as a Green's function to synthesize ground motion of

a large event. The ground motions on the

seismic bedrock at Kannonzaki generated from

sub-faults are produced to match observed

ground motions in a stochastic sense and

superimposed considering time delay due to the

fault rupture process and the wave propagation.

The superimposed motion multiplied by the

transfer function of the ground structure between the ground surface and the seismic

bedrock makes the synthetic motion on the ground surface at Kannonzaki. Table 3 and

Figure 12 show the ground structure model and

226

the transfer function at Kannonzaki used in the

following simulation.

Boorc (1983) modeled the acceleration spectrum

A{f) of shear waves at a site of which distance

R from a source with seismic moment Mo as:

A{f)

F

M. {2nfy

^ KfR ^ R

exp

(3)

where.

F: Radiation pattern coefficient

P- Density at the source fault Shear velocity' at the source fault

fc-. Corner frequency

fmax Cut-off frequency

s : Decay rate at high frequencies

Q{f): Quality factor

Pi, Density at the site A: Shear velocity at the site

Equation (3) was formed following the theories

of source spectrum and the parameters fc, fmax,

and s are chosen to match ground motion

records. The last term of the right-hand side of

Equation (3) is added to consider the

amplification due to the ground structure between the source and the site (Dan et al., 2000).

We employ the shaping window by Jennings et

al. (1968) to obtain the phase spectrum and

generate a transient accelerogram whose

amplitude spectrum matches Equation (3). The

shaping window consists of three parts:

amplifying part (quadratic curve), strong motion

part (constant), and attenuating part (exponential

curve) as shown in Figure 13.
VARIATION OF GROUND MOTION DUE

TO UNCERTAINTY OF SOURCE

PARAMETERS

The effects of the locations of hypoceiiter and

asperities on the simulated ground motions are examined by changing them systematically on

the fault plane.

Variation of Ground Motion due to Location

of Hypoccnter

The ground motions are simulated using 15 locations of hypocenter shown as solid circles in Figure 14. The slip of each sub-fault is assigned

following the variable-slip model of the 1923

Kanto Earthquake (Wald and Somerville, 1995)

shown in Figure 7.

Figure 1 5 shows the acceleration response

spectra (/7=0.05) of the simulated ground

motions. We see the acceleration response varies

about 3 times due to the location of hypocenter

and the variation in long natural period range is somewhat larger than in short period range. The result from the simulation in which the

location of hypocenter of the 1923 Kanto

Earthquake shows the acceleration response

close to the upper bound of the variation in short

period range and mean of the variation in long

period range.

Identification and Simplification of Asperities

Somerville et al. (1999) proposed a procedure to identify rectangle asperities from a variable-slip

model and applied it to 15 crustal earthquakes.

Figure 1 5 shows the rectangle asperities

identified from the variable-slip model of the

1923 Kanto Earthquake by the procedure.

Since two asperities were identified here, we

call the larger one "Asp#l" and the smaller

"Asp//2". The average slip over Asp//1 and

Asp#2 become 2.00 times and 2.09 times of the

average slip over the entire fault, i.e., 2.40[m].

Those values are very close to the value, 2.01,

which Somerville et al. (1999) found as the

average of the I 5 crustal earthquakes.

Therefore, we siiriplify Asp#l and Asp#2 to have uniform slip 4.80[m], Just 2 times of the

average slip over the entire fault. The rest of the

entire fault is set to have uniform slip 1 .57[m] to maintain the average slip over the entire fault. Here we call the simplified slip model

"rectangle asperity model".

Figure 16 compares the simulated ground

motions and their acceleration response spectra

due to the rectangle asperity and variable-slip

models. The location of hypocenter of the 1923

Kanto Earthquake is used here. Since the results show a fairly good agreement, we can use the

rectangle asperity model and change the

locations of Asp#l and Asp//2 to examine the

effects of variation of the slip distribution on the

ground motion.

Variation of Ground Motion due to Location

of Asperities

According to the result of a preliminary analysis,

it was found that the effects of changing the

location of Asp#l in the dip direction on the

ground motion are negligible. Thus, as shown in Figure 17, the locations of Asp#l and Asp#2 are changed systematically with the location of

AspU] in the dip direction is fixed. The location

of hypocenter of the 1923 Kanto Earthquake is used here.

Figure 18 compares the acceleration response

spectra (/7=0.05) from the 18 rectangle asperity

models shown in Figure 17. Figure 18 also

shows the acceleration response spectrum from

the variable-slip model of the 1923 Kanto

227
Earthquake, whicli is almost upper bound of tlie variation of the other 18 cases. This means the

sub-fauhs that liave large shps were

concentrated around Kannonzaki as we can see

in Figure 14. We sec the acceleration response

varies about 3 times and the variation in short

period range is somewhat larger than in long

period range, which is inverse of the effects of

the location of hypocenter.

CONCLUSIONS

Ground motion at the mouth of Tokyo Bay from

tiie large-scale scenario eaithquakes were

synthesized by semi-empirical and stochastic

Green's function techniques. The geometric

source parameters of the 1923 Kanto Earthquake were incorporated into the simulation, while the dynamic parameters, such as fault rupture mode,

locations of hypocenter and asperities were

systematically changed to study their influence on the synthesized ground motion. The

following conclusions may be deduced from the

present study.

1) The peak amplitude of 5% damped

acceleration response spectrum calculated

from the synthesized ground motion for the

Kanto Earthquake is generally about 3000

cm/s^.

2) Variation of fault rupture mode and asperity

affects synthesized ground motion over wide period range, and the spectral

amplitude of resultant ground motion

synthesized by the semi-empirical Green's

function technique changes 3 to 6 times due

to this variation.

3) Acceleration response spectrum estimated

by the stochastic Green's function technique

also varies about 3 times over the natural period range under consideration.

4) The variable-slip rupture model yields

almost upper bound of synthesized ground \ motions within the scope of present study.

This is attributed to a fact that the sub-faults 1 with large slips were concentrated beneath

the ground motion estimation point. '.,

'

1 REFERENCES 1

! V

Boore, D. M. (1983). "Stochastic Simulation of

High-frequency Ground Motions based on

Seismological Models of the Radiated *

Spectra", Bulletin of Seismological Society

ofAmerica, Vol.73, No. 6. Dan, K. et al. (2000). "Isoseismal Map of Strong

Motions for the 1923 Kanto Earthquake ! {MjMA 7.9) by Stochastic Green's Function

Method", Journal of Structural and

Construction Engineering, Architectural

Institute of Japan (in Japanese, Submitting).

Hartzell, S. (1978). "Earthquake Aftershocks as Green's Functions", Geophysical Research

Letters, Vo\.5, No. \. \ Harvard Universit)' (1999). The Harvard CMT i Catalog.

Honshu-Shikoku Bridge Authority (1988).

Seismic Design Guideline of Akashi Kaikyo

Bridge (Draft) (in Japanese). !

Irikura, K. (1983). "Semi-Empirical Estimation

of Strong Ground Motions during Large ! Earthquakes", Bulletin of Disaster

Prevention Research Institute, Kyoto

University, Vol.33, No.298.

Irikura, K. (1986). "Prediction of Strong

Acceleration Motions Using Empirical

Green's Function", Proc. of 7th Japan

Earthquake Engineering Symposium.

Jennings, P. C. et al. (1968). "Simulated

Earthquake Motions", Report of Earthquake

Engineering Research Laboratory, , California Institute of Technology. 1 Kamae, K. et al. (1991). "Prediction of Strong

Ground Motion Based on Scaling Law of
Earthquakes by Stochastic Synthesis

Method", Journal of Structural and

Construction Engineering, Architectural

Institute of Japan, No.430 (in Japanese).

National Land Agency (1998). The Grand

Designfor National Development in the list

Century (in Japanese).

Public Works Research Institute (1991). "Study on Structural Design of Trans-Tokyo Bay

Highway", Report of PWRf Vol.184 (in

Japanese).

Somerville, P. et al. (1999). "Characterizing

Crustal Earthquake Slip Models for the

Prediction of Strong Ground Motion",

Seisniological Research Letters, Vol.70,

No.l.

Taniura, K. et al. (1999). "Developing Seismic

Design Ground Motions for Super

Long-span Bridges", Proc. of 15th U.S.- Japan Bridge Engineering Workshop,

U.S. -Japan Panel on Wind and Seismic

Effects, UJNR.

Wald, D. and Somerville, P. (1995).

"Variable-Slip Model of the Great 1923

Kanto, Japan Earthquake: Geodetic and

Body-Waveform Analysis", Bulletin of

Seisniological Society of America, Vol.85,

No.l.

Table 1 Source Parameters of the 1923 Kanto Earthquake

Fault Length 130 [km]

Fault Width 70 [km]

Strike 290 [deg]

Dip 25 [deg]

Seismic Moment 7.8x10^' [dyne cm]

Rupture Velocity 3.0 [km/s]

Rise Time 5[s]

Table 2 Number of Reassigned Simulation Cases

Rupture starts from the center (l+3)x(3+4)=28 cases Rupture does not start from the center (2+6) x(6+6)=96 cases

Total 124 cases

Table 3 Ground Structure Model at Kannonzaki

Depth

[m]

Density

[t/m']

S Wave Velocity

[km/s]

ground surface 0 31 1.9 0.42

31 45 1.9 0.40

45 57 1.9 0.50

57 77 1.9 0.61 77 94 1.9 0.70

94 120 1.9 0.84

120 1005 2.0 1.15 1005 1405 2.0 1.20 1405 2605 2.1 1.30 2605 3005 2.3 1.40

seismic bedrock 3005 - 2.5 2.72

229
4^

Time

Figure 1 Source Time Function (Irikura, 1986)

J s Epicenter of Kanto Eq

Epicenter of

Cfi|ba-ker

Kannpnzaki

o

Tolio-oki Eq,

Figure 2 Fault Plane, and Locations of Epicenters and Kannonzaki

Figure 3

15 20 25 30 35 40

Time (s) Kannonzaki Record from the 1987 Chiba-ken Toho-oki Earthquake

(N-S Component)

230
o

ft

CD

O• r—

(

o

o

<

0000

000

h=0.05

.00

0

0.1 1 5

Natural Period [sec]

Figure 4 Acceleration Response Spectrum of Kannonzaki Record (/7=0.05;

(U

SE

NW SE

NW

a.

T3

SE NW

(a) Unilateral and Bilateral Rupture

NW

NW

-a

SE NW

(b) Hypocenters of Radial Rupture

Figure 5 Rupture Modes and Hypocenters of Radial Rupture

231
NW SI-, NW

NW SI-: NW

NW Sl£ NW SE

(a) 2x3 Asperity

NW

SE NW SE NW SE NW

SE NW SE NW SE NW

SE NW SE NW SE NW

(b) 3x2 Asperity

Figure 6 Asperity Shape and Locations

B 20

Slip Vectors

1 I -V1

-I- 1

1 1 \ 4-

1

ir \ -> —

\ \

> \ V N \ ^ \

' \ -» > \

\ \ X J

-> >

1 I

1 V

1

J

1 120 100 80 60 40 20 0 NW along strike fault length (km)

10m

SE

Figure 7 Variable-Slip Rupture Model of the 1923 Kanto Earthquake (Wald and Somerville, 1995)

232
NW SE NW SE

(a) Unilateral and Bilateral Rupture

NW

NW

(b) Hypocenters of Radial Rupture

Figure 8 Reassigned Rupture Modes and Hypocenters of Radial Rupture

SE

SE

NW SE NW SE

NW SE NW SE

(a) 2x3 Asperity

NW

NW

SE

SE

NW SE NW SE

NW SE NW

(b) 3x2 Asperity

SE

s NW

NW

Figure 9 Reassigned Asperity Shape and Locations

233
CDO

0.1 1 5 Natural Period [sec]

Figure 10 Acceleration Response Spectra of Synthesized Ground Motion

: 0.1 15

Natural Period [sec]

Figure 1 1 Acceleration Response Spectrum using the Variable-Slip Rupture Model

234
10 I

'

'

'

I

1

^

^

1 -

0.5 ' ' ' ' 0.05 0.1 0.2 0.5 1 2 5 10

Period [s]

Figure 12 Transfer Fuiiclion of the Ground Structure between the Ground Surface and the Seismic

Bedrock at Kannonzaki

amplify ing part

Figure 13 Shaping Window (Jennings et al., 1968)

235
139° E 140° E

km

0 50 -

, 100 hypocenter of the

1 923 Kanto earthquake

35° N

0 2 4 6 8

slip [m]

Figure 14 Rectangle Asperities Identified from the Variable-Slip Model of the 1923 Kanto Earthquake

and the Locations of Hypocenters used in the Simulation

w 5000

w

E

•I 1000

03

0)o

o

cd

CDw

c

o

Q.

(0CD

500

100

50

0.1 1 period (s)

Hypocenter of the 1923 Kanto

earthquake

Figure 1 5 Variation of Ground Motion due to Location of Hypocenter

236
" -1200 I ' • ' ' ^ ' 10 20 30 40 50 60 70

lime (s)

" -1200

1

1 Variable-slip model

10 20 30 40 50 60 70

time (s) (a) Waveforms

5000

Variable-slip model

0.1 1 period (s)

(b) Acceleration Response Spectra

Figure 16 Comparison between the Simulated Ground Motions due to the Rectangle Asperity and

Variable-slip Models

237
Asp#2 Asp#1

Hypocenterr o Kannonzaki

o

Figure 17 Rectangular Asperity Models with Variety of Locations of Asp#l and Asp#2

The locations of hypocenter and Kannonzaki are also shown.

w 5000

E

o

03

i_

CDOO

03

CDCO

£=O

Q.

COCD

Variable-slip model

0.1 1 period (s)

Figure 18 Variation of Ground Motion due to Location of Asperities

238
DEVELOPMENT OF SEISMIC ISOLATION TECHNOLOGY

FOR UNDERGROUND STRUCTURES AND THE APPLICATION

by

Shigeki UNJOH", Junichi HOSHIKUMA", Kazuhiro NAGAYA" and Takeyasu SUZUKI''

ABSTRACT

The seismic isolation for underground structures

is the system in which a turmel body is covered

by a thin isolation layer consisting of elastic materials with low shear modulus. The isolation

layer absorbs the strain transmitted from the surrounding soils during large earthquakes so

that the seismic performance of underground

structures be improved. The Public Works

Research Institute has developed this new

seismic isolation technology with the Public

Works Research Center and 17 private

companies from 1995-1998. Then the seismic

isolation technology is applied for a practical

constrction at the connection of vertical shaft and tunnel of common utility ducts firstly in the world. This paper presents the outline of the seismic isolation technology and the practical

application.

KEY WORDS: Seismic Isolation

Underground Structures Seismic Performance

Common Utility Ducts

1. INTRODUCTION

The Hyogo-ken Nanbu Earthquake of January

1995 damaged underground structures and some

underground structures such as the collapse of

the Daikai Station on the Kobe Rapid Transit

Railway were heavily damaged. It is essential to guarantee the operation of communication,

electricity, gas, water, and other lifelines in regions struck by powerful earthquakes. To

achieve this goal, the safety from strong

earthquakes of utility turmels that are typical

undergroimd structures provided to concentrate

these lifeline services is required to be

preserved. And while the earthquake resistance

of underground structures must be improved,

the public is demanding that the cost of their

construction be reduced.

To meet these needs, the Public Works

Research Institute of the Ministry of

Construction, the Public Works Research Center

and 17 private companies have taken part in joint research to develop seismic isolation technology for underground structures as a new

technology that can improve the earthquake

resistance of underground structures over that provided by conventional technologies at the same tune as it lowers construction costs

"

The seismic isolation structure that was

developed for underground structures is a

structure that can sharply reduce the effects of

earthquakes by forming a flexible seismic

isolation layer around the outer periphery of an underground structure to insulate the

underground structure from deformation of the surrounding ground.

This paper describes an outline of seismic

isolation technology for underground structures and a practical application for the starting vertical shaft of the "Nakagawa Utility Tunnel.

An outline of the test construction of the seismic isolation structure as well as the field

test to verify the seismic isolation functions are presented.

1) Head, Earthquake Engineering Division,

Earthquake Disaster Prevention Research

Center, Public Works Research Institute,

Ministry of Construction, Tsukuba-shi,

Ibaraki-ken, 305-0804, Japan

2) Senior Research Engineer, ditto 3) Research Engineer, ditto 4) Deputy Director, Techical Institute of

Kumagai-gumi Corp., Tsukuba-shi,

Ibarakli-ken, 300-2651, Japan (Underground

Structure Seismic Isolation Technology

Promotion Committee)

239
2. R&D OF SEISMIC ISOLATION

TECHNOLOGY

Fig.l shows a conceptual diagram of the seismic isolation structure. Seismic isolation

mechanisms for underground structures are categorized under the following four items.

1) Blocking the transmission of strain during

earthquakes from the surrounding ground to a turmel structure 2) Smoothing the tunnel strain in the

seismic-isolated structure ( longitudinal

direction)

3) Dispersing the stress resultant concentrated at the corners of rectangular section tunnels

(lateral direction)

4) Insulating the vertical shaft body from the tuimel at the turmel entrance and smoothing the tunnel strain in the seismic isolated section (at vertical shaft coimection points)

The performance required by a seismic isolation

layer capable of providing these mechanisms was studied from the perspective of its mechanical properties and construction

properties, then the seismic isolation material was developed based on the results of the study.

The development process included testing to corroborate the mechanical properties,

durability, etc. of the seismic isolation material,

testing to verify its construction properties by

trial manufacturing a seismic isolation material

injection system, and corroborative material

injection trials in an actual turmel. And to develop a seismic isolation structure

construction method for shield turmels, in addition to a seismic isolation structure for use

at points where ground conditions change

abruptly, a seismic isolation structure for

vertical shaft connections that are points of

abrupt change in the structvu-e and a method of

constructing such a seismic isolation structure have been proposed. A design method and

construction method for seismic isolation

structures for undergroimd structures that has

been established as a result of this joint research were published in September 1998 as the Underground Structure Seismic Isolation Design

Method Manual (Draft)

The joint research has clearly shown that the seismic isolation structure can be effectively

applied in l)the longitudinal direction at points

of abrupt change in ground conditions, 2) where

the lateral section of a rectangular tunnel

changes abruptly, and 3) in the longitudinal

direction at connections with vertical shafts, but

the results of a cost-effect analysis show that the

most promising application of this seismic

isolation structure is at the connections with

vertical shafts of a shield tunnel.

The conventional method of providing seismic

protection at connection with vertical shafts of a

shield tunnel has been the application of a

flexible segment. A flexible segment is a single

special segment ring constructed so that it concentrates and absorbs strain generated in the surrounding tunnel. A flexible segment can

absorb tunnel strain that is locally concentrated

at the location where it is installed, but its effects are generally limited to a small range. But if a vertical shaft is isolated from a segment

by a seismic isolation layer at the same time as

a seismic isolation layer is installed within the range where the ground strain is most

concentrated for about 10 m from the shaft, it is possible to provide isolation from both a point

of abrupt change in the structure and from

concentrated ground strain. In this way, the turmel stress resultant can be reduced in a short seismic isolation section .

3. PRACTICAL APPLICATION OF

DEVELOPED SEISMIC ISOLATION

TECHNOLOGY

3.1 Outline

The seismic isolation structure for underground

structures has progressed to the research and

development stage, but because it has still not reached the stage where it is a practical design

and execution technology, an actual structtire must be designed and a trial execution of this design performed;

In 1997, initial seismic design for LI earthquake

motion was performed for the vertical shaft cormection point of the No. 1 Nakagawa Utility

Tunnel (below referred to as "Nakagawa Utility

Tunnel") , and a flexible segment was chosen

for the start side and a flexible joint for the

arrival side of the connection point. The Nagoya

240
National Highway Construction Office of the Chubu Regional Construction Bureau of the Ministry of Construction applied the field trial system, the connection was redesigned to provide seismic resistance capable of

withstanding L2 earthquake motion one ranJc higher than the previous design, and to confirm

the applicability of the new system in actual tunnels and execution costs, this seismic

isolation technology was selected for execution on the No. 1 Nakagawa Tunnel in 1997. This was the first attempt to use this seismic isolation

structure anywhere in the world.

3.2 Seismic Isolation Design of Vertical Shaft

Connection

(1) Structure Conditions

Fig.2 shows the surface ground structure at the

start side of the vertical shaft and the locational

relationship of the vertical shaft and tunnel. The

bottom plate of the vertical shaft is in a diluvial

clay layer, and the SMW wall diaphragm that surrounds it extends for about 40m down to a

diluvial sandy soil layer. The earthquake wave

input bedrock and the bedrock of the

axisymmetric finite element method model for

this design were assumed to be the top of the

diluvial sandy soil layer Dss (V< = 320m/s) at a depth of 30.9m in the figure. The figure also shows the mechanical physical properties during

slight strain of the surface ground.

The vertical shaft is a reinforced concrete

structure with external width of 14m and depth

of 10.8m and the plane shape shown at the top

of Fig.2. Its wall thickness varies somewhat

according to the depth and it has partition walls,

but because the overall shear stiffness is not

effected by accounting for these, it was assumed

to be a box-shaped structure with a constant wall thickness of 14m for this study. The

modeling of the vertical shaft by an

axisymmetric finite element method model

included the modeling of not only the vertical shaft body, but also the wall diaphragm

constructed by SMW as part of the vertical

shaft.

The tunnel studied was a shield tunnel with an

exterior diameter of 5,050mm made of

reinforced concrete segments with a thickness of

250mm, it was constructed by the slurry shield

method ( external diameter of the shield machine: 5,190mm) , and its segment ring joints were connected with long bolts. The tension

stiffness of the shield tunnel was modeled using

equivalent tension stiffness calculated from the spring of the segments and from the serial spring of the spring of the ring joints. The

compression stiffness used was the compression

stiffness of the concrete of the segments.

Because the external diameter of the shield

machine was 5,190mm, the thickness of the

isolation layer was set at 7cm that equals the thickness of the tail void.

The verification of safety during an earthquake

of the shield tunnel was performed based on two factors the allowed aperture of the ring joint

(2mm) and the yield stress of the joint bolts (940N/mm') during tension deformation, and

based on the allowed compression stress

( 24N/mmO of the segments during

compression deformation.

(2) Input Earthquake Motion

The input earthquake motions for L2 earthquake

motion were chosen from among standard waves used for seismic design of highway

bridges. Waves obtained by performing

amplitude compensation of the component at right angles to the bridge axis of the Kaihoku

Bridge during the Miyagi Prefecture Offshore

Earthquake of 1978 were used as Type-1

earthquake motion for category I ground. Waves

obtained by performing amplitude compensation

of the NS component recorded at the Kobe

Marine Meteorological Observation Station

during the Hyogo-ken Nanbu Earthquake of

1995 were used as the Type-2 earthquake

motion. Fig3(l) shows the Type-1 earthquake

motion and Fig.3(2) shows the Type-2

earthquake motion. The one-dimensional seismic

response analysis of the surface ground was

performed by the substitute structure approach:

considering these input earthquake motions to be waves observed at the outcrop of fi"ee rock,

reducing the amplitude by 50%, and inputting the resulting waves fi-om the top of the bedrock

layer D.3. The strain dependency of the ground

used for the analysis was obtained from a mean

strain dependency evaluation equation for

typical ground " '\

241
The ground inertial force applied by the axisymmetric finite element method model was

calculated by the ground response seismic

coefficient method" that uses the ground

acceleration at the time when the maximum

value of the total of the shear strain of the ground from the bed rock to the surface is generated based on seismic response analysis.

(3) Design Results

Fig.4 shows the results of a parametric study of

deformation during tension subject to strict design conditions that was conducted to determine the shear modulus of rigidity of the seismic isolation layer and the length of the seismic isolation section. The variables for the parametric study were shear modulus of rigidity

of the isolation layer of 0.3 and 0.5N/mm" and

the length of the seismic isolation section from

the exterior wall of the vertical shaft that were

10, 15, and 20m. The shearing modulus was set considering silicon seismic isolation material

(SISMO-003 or SISMO-005) that is isolation layer material with superior durability and

workability. Analysis results have revealed that within a seismically isolated section length

between 10m and 20m, there is little variation

in the tension axial force of the tunnel and that even if the shearing modulus of the seismic

isolation layer is reduced from 0.5N/mm^ to 0.3N/min\ the tension axial force of the tunnel

declines very little. The shearing modulus of the seismic isolation layer was, therefore, set at 0.5N/nim^ and the seismically isolated section

length was set at 10m from the exterior wall of

the SMW.

Fig.5 shows an example of the results of an

analysis of the joint aperture during tension

deformation caused by Type-1 earthquake

motion. In addition to showing the results for a seismic isolation structure case, the figure

presents the results of an analysis of a rigid

connection case and of a case hypothesizing a

flexible segment with a Young's modulus of

O.lN/mm" at a point 1 m from the external wall

of the vertical shaft. If a seismic isolation structure is adopted in this way, the ring joint

aperture will be within the allowed value

(2mm) , but in the cases hypothesizing the adoption of a rigid connection and a flexible

segment, the aperture exceeded the allowed

value. The tension stress of the joint bolts was

within the range of the allowed values for all cases.

Fig.6 shows the results of an analysis during

compressive deformation. This figure shows that under both types of earthquake motion, the compressive stress of the segment was within

the allowed value (940N/m") for all cases. It also reveals that adopting the seismic isolation structure lowers the compressive axial force on

the segment to about half that obtained with the

rigid connection method, which is larger than

the reduction to 2/3 possible with a flexible segment.

Fig.7 is a schematic diagram of the seismic

isolation structure at the start vertical shaft established by the above seismic isolation design process. The design calculations modeled

the SMW and the vertical shaft body as a

vertical shaft. The seismic isolation section that was analyzed extended for 10m from the external wall of the vertical shaft, but for execution reasons including the fact that the segment length is 1.2m, the seismic isolation layer was formed in the tail void for 10 rings

from the tunnel entrance. This means that the seismic isolation section extended 9.6m from

the external wall of the SMW and 10. 5m from

the external wall of the vertical shaft body. On

the ground side of the SMW wall, the seismic

isolation layer was 7crn thick that is the same as the tail void, but it was thicker at the entrance. And to provide the tunnel entrance with

waterproof properties and to insulate the vertical shaft body from the segment, a natural rubber

ring with a wedge-shaped section, covered on

three sides by water-absorbing rubber, and with

a shearing modulus of 0.6N/nim^ (tunnel entrance insulation layer) was installed at the tunnel entrance and integrated with the tunnel entrance concrete.

The adoption of a seismic isolation structure as described above, provides an existing structure with the ability to withstand Type L2

earthquake motion. And the waterproofing

function of the silicon seismic isolation material

guarantees waterproofing even if an

unpredictable joint aperture occurs.

242
3.3 Practical Construction

(1) Isolation Material

The seismic isolation material used for the trial execution was silicon seismic isolation material

SISMO-005 with its shearing modulus G

adjusted to 0.5N/mm" or less. Silicon type

seismic isolation material is a rubber elastic body hardened at normal temperatures by the condensation reaction caused by mixing and

agitating fluid A that is a mixture of silicon polymer + finely powdered silica and fly ash

with fluid B that consists primarily of a hardener and plasticity adjustment agent at a mix ratio by weight of 10:1. A cross linking

agent is added uniformly to fluid A and a catalysis is added to fluid B. It is, therefore,

possible to control only the hardening time

without altering the hardness of the rubber that

is the seismic isolation material by adjusting the quantity of the catalyst in fluid B. The seismic

isolation material is an inorganic material based on silica and is both non-polluting and durable. Fig.8 shows the strain dependency of the shearing modulus of the seismic isolation material obtained by cyclic hollow cylindrical

cyclic shear testing performed at a loading

frequency of IHz. This test shows that the seismic isolation material has stable dynamic

physical properties regardless of the shearing

strain. The material has no constraining pressure

(depth) dependency and it is an elastic material

with a damping factor less than 3%'\

Fluid A is supplied to the agitator (capacity

1.8m') of the plant from a hopper located

above the ground, then it is pressure fed from

the agitator at a fixed rate by a mono-pmnp.

Fluid B is supplied to the fluid B agitator

(capacity 0.2m') of the plant from, a drum, then

it is pressure fed from the agitator at a constant

rate by a mono-pump. The tip of the injector is a rubber injector used for backfill injection.

After the two fluids have been mixed, a quarrying column mixer is passed through a 2-inch steel pipe for a total of 80 cm and its pressure loss is utilized to completely mix the

fluids. Because this mixer is long and it is difficult to work with, the injector unit and

mixer are connected by a flexible rubber tube.

The specific gravity of the seismic isolation

material is 1.36, and its elongation is 100%. As

stated above, the mix proportion was designed

to lower the shearing modulus to less than 0.5 N/mm". Therefore, when the seismic isolation material is shipped from the factory, it provides

JIS hardness of 32 degrees and a shearing

modulus of 0.44N/mm' and satisfies the required physical properties after it has been

blended indoors based on the stipulated mix

proportion and is cured for one week at a temperature of 20 Centi-Degree.

(2) Construction

The level difference between the injection pump

and injection oudet was minimized, care was

taken to obtain a stable blend of fluid A and

fluid B, and the seismic isolation material plant was installed in an intermediate level of the

vertical shaft. Fig.9 is a schematic diagram of

the seismic isolation structure at the start side of

the vertical shaft. As shown in the figure, the seismic isolation material was, in principle,

injected during excavation from injection holes

in the segments into the tail void produced as the shield excavator advanced in the same way

as immediate backfilling injection is performed.

It was feared that during the seismic isolation material injection work, seismic isolation

material might flow into the blades where it would get into the chamber. To prevent this from occurring, as the shield excavator

advanced, void filling clay (clay shock) was

injected into the void between its steel shell and

the natural ground from an injection outlet

installed near the articulation of the shield excavator to block the gap between the tail void

and the blade. Because the clay shock was

pressure injected into small voids on the excavated natural ground, it also prevented the seismic isolation material from leaking into the

natural ground.

At locations in the fluid A and fluid B

pressurized feed hoses just before their respective injection and mixing locations,

bypasses were installed to cycle the two fluids through agitators prior to injection in order to prevent separation of their materials and a decline in their viscosity. To form a seismic

isolation layer of uniform thickness around the segments, it was vital to exercise strict control of the excavation. It was necessary to guarantee

243
that the seismic isolation work would not

conflict with the basic principles of shield work

that require the immediate filling of the tail void

to prevent the collapse of the natural ground and

to keep out ground water and mud. To

guarantee this, before the work began, the segments were equipped with injection holes at three locations in the tunnel axis direction that differs for each segment piece, and injection was begun from the injection hole closest to the void, to permit the restarting of excavation after

it was stopped because of trouble with the work.

That part of the work to form the seismic

isolation structure at the vertical shaft

connection that was most important and had to be done with the greatest care was the work at the entrance: the first stage in the execution.

Because a large void is formed behind the segment at the entrance as shown in Fig.9, it was first stabilized by filling it with high

density mud. After excavation to the third ring,

clay shock was injected to fill the void behind

the steel shell of the shield excavator to prevent

the seismic isolation material from flowing into the blades, then the injection of the seismic

isolation material at the entrance was performed.

(3) Quality Control

To control the quality of the seismic isolation material, the hardness of a hardened sample of

the seismic isolation material made at the time it was shipped from the factory and the hardness

of hardened samples made before, during, and

after injection were measured by the

JIS-K6301-A method. The flow rates of the seismic isolation fluid A and fluid B were

controlled by installing Coriolis flow meters,

and the quantities of fluid A and fluid B

remaining in their respective tanks after the

injection work were measured to confirm that the fluids were completely injected.

While the mean hardness of the seismic

isolation material when it was shipped from the factory was 32, the hardness of samples of

seismic isolation material obtained at the tip of

the injector immediately before, during, and

after injection and cured for 1 week ranged

from 26 to 33 and the mean value was 29. Converting this to the shearing modulus

obtained a result of 0.39N/mm\ which is a little

lower than the results of the measurements at the factory (hardness of 32, shearing modulus

0.45N/mm') . The mean hardness of the seismic

isolation material after 1 month passed, was 31

(G = 0.42N/mmO, a result that fully satisfied the design value of G = 0.5N/mm' or less.

(4) Field Test for Effectivenes Verification

To verify the seismic isolation performance of

the seismic isolation structure formed at the

vertical shaft coimection by this trial execution,

a field experiment was carried out to measure

the shear reaction force in the tunnel axis

direction of the seismic isolation layer. When a

seismically isolated shield tunnel sunounded by

a seismic isolation layer is forcefully deformed

in the tunnel axis direction, shear resistance

proportional to the shear reaction of the periphery of the turmel is obtained. Because this shear reaction force is the shear resistance of the seismic isolation layer and the ground around it,

it is possible to evaluate it based on their series spring system. If the shearing modulus of the seismic isolation layer is more than 2 orders

smaller than that of the ground, it is possible to approximate the shear reaction force of the seismic isolation layer body that treats the ground as a rigid body. Consequently, using the reaction force of the wall of the vertical shaft,

the tunnel was moved towards the cutting face

along with the shield excavator by a hydraulic

jack, and the shear spring constant of the

seismic isolation layer was estimated from the relationship of the' tunnel axial direction

deformation with the thrust of the hydraulic

jack.

The shear spring constant of the seismic

isolation layer can be obtained theoretically by

Eq.2 from the relationship between the force

and the displacement produced by treating the

periphery of the seismic isolation layer as a

constant condition and forcefully deforming the external surface of the txmnel in the turmel axial

direction.

K. = 2K *G-/ln (Rn/R.) * L (l)

Where:

Kmi shear spring constant of the seismic

isolation layer

Gr,: shear modulus of the seismic isolation layer L: length of the seismic isolation layer

244
1/2 of the external diameter of the seismic isolation layer

Rr. Ill of the external diameter of the tunnel

Because it was necessary to complete the

injection of the seismic isolation layer at the

starting vertical shaft connection to form the seismic isolation then perform this test before

the backfill injection that accompanies normal

shield excavation, the test was performed four

days after the final day of the seismic isolation material injection work.

Fig.lO is a schematic diagram of the seismic

isolation layer reaction force measurement test. The reaction receiver installed on the vertical shaft body wall was a mobile structure that allowed the removal of the bolts to cause the temporary segment reaction frame to slide towards the tunnel. The reaction receiver was

equipped with a pedestal to be used to install eight hydraulic jacks. By removing its bolts and

installing jacks on the pedestal, the reaction

receiver becomes a loading device structurally

equipped v^th a movable reaction frame. As

stated above, the periphery of the shield excavator is filled with clay shock and it is assumed that the shear resistance of the periphery of the shield excavator is low. It was,

therefore, possible to move the segment in the direction of the cutting face along with the

shield excavator to the location where the shear

reaction force of the seismic isolation layer and

the jack thrust were in balance, by advancing

the hydraulic jacks installed on the reaction

receiver and at the same time, adjusting the

cutting face slurry pressure downwards to prevent the action of the jacks Jfrom increasing

this pressure to maintain it at a constant level.

The jacking pressure was increased at steps of

50kN steps per jack for a total of 400kN per

step. The jacking pressure was increased again

after the tunnel deformation caused by the previous increase in thrust had stabilized. The

loading was concluded when the turmel had

been deformed 10mm towards the cutting face. The unloading was performed in the same way

as the loading: reducing the pressure applied by

the jacks by 400kN, waiting for the tunnel

deformation to stabilize, then lowering the

pressure by another 400kN step. When the hydraulic jack pressure was down to zero, the residual displacement of the tunnel was

measured. The jacks were operated from an

intermediate level in the vertical shaft at a location where the operator could see the loading device equipped with the movable

reaction frame.

The cutting face slurry pressure was

automatically regulated so that it remained at a constant level of O.lSMPa (reaction of 3,200kN

to by the slurry pressure) throughout all loading and unloading steps. But during the loading steps, the cutting face slurry pressure temporarily rose to a maximum of 0.17MPa and

it took time to lower it to its stable value. But

during unloading, the cutting face slurry pressure remained stable at almost exactly

0.15MPa, with the resuU that the pulling phase

of the loading operation did not take as long as the pushing phase.

The measurements were performed by

monitoring the pressure and stroke of the 8

hydraulic jacks, the relative axial direction

displacement of the segment and the vertical shaft body on the left and right sides of the

vertical shaft entrance, and the cutting face

slurry pressure in the central control room, and

by measuring the distance between the vertical shaft and the ring 1 joint of the segment closest

to the cutting face using a light measure system

fixed at the vertical shaft portal. The data

obtained from these measurements was

displayed to the operator in the intermediate

level of the vertical shaft and used to make jack

operating decisions. It was also monitored and

recorded in a measurement vehicle located at ground level.

(4) Test Results

Fig.ll summarizes the results of the test in terms of the relationship of the relative axial direction displacement of the turmel and the

vertical shaft body wall at the entrance with the

total thrust Pt of the 8 jacks installed on the reaction frame. The hysteresis in this figure is categorized as 5 loading states from I to V to explain the process of determining the shear

spring constant of the seismic isolation layer.

While the behavior varied according to the

245
loading stage as shown in Fig. 11, the tunnel

stopped at location F with residual displacement

of approximately 1.57mm without being

restored to its original position even after the jack pressure was finally reduced to zero at the end of the unloading steps.

With the residual displacement of the tunnel at location F represented by 6 cm, the shear

spring constant of the seismic isolation layer by

Kr, (kN/cm) , the maximum static friction force

of the temporary segment and the pedestal by

sialic, and the cutting face pressure produced by

the cutting face slurry pressure by Pmud, and

assuming that the cutting face slurry pressure can be ignored because in static state, it is balanced with the slurry pressure, the following

equation of equilibrium can be established for the point in time where the tunnel has stopped

at the final stage.

Km = 5 + Pmud = jPjMfic (2)

It can be concluded that because the residual

displacement (5 = 0.157cm, the maximum static friction force is in balance with (reaction force

produced by the cutting face slurry pressure = 3,200kN) + (reaction force of the seismic

isolation layer equivalent to the shear

deformation of 0.157mm) . Next, if it is assumed that the point A in Fig.ll corresponds

to the time when the temporary segment began

to slide on the top of the pedestal, the maximum

static friction force Pstatic — 4,570kN, and if these

values are substituted in equation (2),

(kN/cm) is as shown below. = 4,570 3,200/0.157 = 8,730 (3)

The thick dark line in Fig.ll shows the results of assuming that Pslalic — 4,750kN and = 8,730kN/cm to estimate the hysteresis

relationship of the jack thrust with the axial displacement of the tuimel when the reaction

force at the cutting face is assumed to be

constant. The hysteresis that was estimated

shows a separation from actual behavior caused

by instability of the cutting face soil pressure at loading stage III but the fact that both

conform closely at loading stage II and at V that was the final stage of the unloading and an

almost linear load displacement relationship was obtained during unloading reveals that a seismic isolation layer with reliable resilient behavior was formed around the periphery of

the tunnel and its spring constant Kn =

8,730kN/cm.

Because the mean injection rate of the seismic

isolation material was approximately 120%, the

seismic isolation layer corresponding to this injection rate is 8cm, and the theoretical value

of the shear spring constant of the seismic

isolation layer with a shear modulus G = 0.5N/mm' is ll,380kN/cm' based on Eq.(l).

Because the measured shear spring constant of

the seismic isolation layer is 8,730kN/cm, the

shear modulus of the seismic isolation layer is 0.383N/mm", which conforms almost perfectly

with the mean value of 0.39N/mm^ of the

seismic isolation material injected as described

above following 1 week of aging. The shear

spring constant and the shearing modulus of the seismic isolation layer are, therefore, appropriate

values.

Because as described above, elastic behavior

capable of restoring the tunnel to its original

position during unloading in a linear relationship

with the jack thrust was obtained and because

the shearing modulus of the seismic isolation

layer is an appropriate value, it has been

verified that a seismic isolation layer with the

intended functions was constructed.

4. CONCLUDING REMARKS

This paper has surnmarized the design and

execution of a seismic isolation structure for underground structures that has been made at the start vertical shaft connection of an actual

shield tunnel.

(1) The trial execution of the seismic isolation structure verified the execution properties of the

silicon seismic isolation material injection

system.

(2) The mean injection rate of the seismic

isolation material is 120%, which is a rate lower

than the injection rate of normal backfill. (3) In-situ testing conducted to measure the

reaction force of the seismic isolation layer

verified the functions of the seismic isolation structure by demonstrating that elastic behavior

capable of restoring the tunnel to its original

position during unloading in a linear relationship

with the jack thrust could be obtained and that

246
the shearing modulus of the seismic isolation layer obtained by the test conforms with

measured values.

(4) The above results have shown that this seismic isolation structure for underground

structures is a technology that can be applied to the design and execution of actual underground

structures.

ACKNOWLEDGEMENTS

The construction of the seismic isolation structure was performed with the generous

assistance and cooperation of the staff of the Nakagawa Shield Tunnel Construction Office,

Hazama-Sato Joint Venture under the contract of Nagoya Construction Office of Ministrry of

Construction. The authors wish to conclude by

expressing their sincere gratitude to them for

their contribution to this project. The members

of the Underground Structure Seismic Isolation Technology Promotion Committee provided

technical advice regarding the planning and the design of the trial execution. The authors wish

to also thank them for their assistance.

REFERENCES

1) Public Works Research Institute, et. al.: Report on Joint Research on the Development

of Seismic Isolation Materials for Use in the Seismic Isolation Design of Underground

Structures (Part l) , Joint Research Reports,

Reference No. 154, Nov. 1996 (in Japanese)

2) Public Works Research Institute, et. al.: Report on Joint Research on the Development

of Seismic Isolation Materials Tor Use in the Seismic Isolation Design of Underground

Structures (Part 2) , Joint Research Reports,

Reference No. 192, Dec. 1997 (in Japanese)

3) Public Works Research Institute, et. al.: Report on Joint Research on the Development

of Seismic Isolation Materials for Use in the Seismic Isolation Design of Underground

Structures (Part 3) , Underground Structure Seismic Isolation Design Manual, Joint

Research Reports, Reference No. 211, Sept.

1998 (in Japanese)

4) Shigeki Unjoh, et. al.: Seismic Isolation Structures and Seismic Isolation Design of

Underground Structures, Proceedings of the Sixth Tunnel Engineering Research Conference,

Vol. 8, pp. 31-38, 1998 (in Japanese)

5) Public Works Research Institute: Numerical

Analysis of Response Properties of Ground

During Earthquakes, - DESTRA -, PWR

Document No. 1778, February 1982 (in

Japanese)

6) Yasuda, Yamaguchi: Dynamic Deformation

Properties of Various Categories of Undisturbed

Soil, Twentieth Soil Engineering Research

Conference, pp. 539-542, June 1985 ( in Japanese)

7) Katayama Ikuo et. al.: Proposal of the "Response Seismic Coefficient Method" a Practical Quasi-dynamic Analysis Method for Structures Buried Underground, Fortieth Aimual

Technical Conference of the Japan Society of

Civil Engineers, Part 1, pp.737-738, 1998 (in

Japanese)

8) Takeyasu Suzuki et. al. Dynamic Properties

of Silicon Seismic Isolation Material for Tunnel

Seismic Isolation Work, Proceedings of the

Thirty-second Conference on Ground

Engineering, pp. 2095-2096, 1997 (in Japanese)

9) Takeyasu Suzuki, et. al.: Experimental Study

of Silicon Material for Urban Tunnel Seismic

Isolation Structures, 'Collected Reports of the

Japan Society of Civil Engineers, No.

534AT-30, pp.69-76, 1996 (in Japanese)

10 Shigeki Unjoh, et. al.: Function Verification

Testing of Seismic Isolation Structures for Use

at Vertical Shaft Connections, Collected Papers

and Reports on Tunnel Engineering Research,

Vol. 9, 1999 (in Japanese)

11) Shigeki Unjoh, et. al.: Draft Manual for Seismic Isolation Design for Underground

Strucutres, Proceedings of the 30th Joint

Meeting on Wind and Seismic Effects, UJNR,

NIST SP931, August 1998

247
Fig.l Seismic Isolation Technology for Underground Structures

0.9ni 15£m U 14.0m

,.4m 1.4m hi-—11.2m -I I

0.9m

SMW Diaphragm Wall /

Vertical'shaft Wall /

//////

CO

SMW Diaphragm Wall

Bedrock for Design Analysis

_J

Fig.2 Layout of the Vertical Shaft and Tunnel

248
1000

500

o -500 <

-1000

If!

10 15 20 Time (sec)

25 30 Time (sec)

(l) Type-1 (2) Type-2

Fig. 3 Input Earthquake Motion

12000

10000

8000

6000

4000

i 2000

0

Rigid Connection

G=0.5N/mni', 10m

- — G =0.3N/mm', 10m

G=0.5N/minM5m

G=0.5N/mni', 20m

J J0 10 20 30 40 50

Distance from External Wall of Vertical Shaft (m)

Fig.4 Effect of Shear Modulus and Length of Seismic Isolation Layer

2.5

2.0

1.5

< 1.0

0.5

0.0

! f\ - T YPE-1

1

1*

1

1 Rigid Connection -

Flexible Segment

Seismic Isolation

'

1

1 f Mi

-1-1

0 50 100

Distance from External Wall of Wertical Shaft (m)

150

Fig.5 Analysis of Ring Joint Aperture

249
100000

S 80000

^ 60000

40000

20000

0

} TYPE-1

1 1

1 r

r

I

-1

- Rigid Connection

Flexible Segment

Seismic Isolation -

1 J

-•»L.. -T

... ^ _

-n..

0 50 100 150

Distance from External Wall of Vertical Shaft (m)

Fig. 6 Analysis of Compressive Axial Force of Segments

V

Styrofoam Block

Seismic Isolation Layer : Backfill Material ^ C,...—

Entrance Insulation Layer

I 1... I Entrance Concrete \77~r7-rzyv

Vertical Shaft ^ V////////.

900

iS9..600-

^i();500^ Ground

Fig.7 Seismic Isolation Structure at Start Vertical Shaft

EE

C5

1.0

0.8

0.6

§ 0.4

to ^ 0.2

0.0

SISMO-5

0.0001 0.001 0.01 0.1 Shear Strain r

Fig. 8 Strain Dependency of Shear Modulus of Silicon Seismic Isolation Material

250
Fig.9 Silicon Seismic Isolation Material Injection

Fig. 10 Schematic of Field Measurement Test

251
Axial Tunnel Displacement (mm)

Fig. 11 Jack Thrust and Axial Tunnel Displacement Relationship

252
EARTHQUAKE MOTIONS AT AN EMBANKMENT DAM BASE

AND AN ESTIMATION METHOD OF INCIDENT SEISMIC WAVES

USING THE OBSERVATIONS

Yoshikazu Yamaguchi'^ & Tomoya Iwashita

ABSTRACT

Many earthquake motions have been observed at the bedrock of embankment dams. The observed waves

are affected by the dynamic interaction between the dam and its foundation. The effects of the frequency

contents of the input waves and the impedance ratio on the dynamic interaction at the dam base were

evaluated using FEM analyses. The change of the dynamic interaction on the footprint of dams was

represented analytically. The analytical results agreed

with the earthquake observations and the miCTOtremor measurements at the dam sites. Moreover, a simplified analytical procedure to estimate incident seismic wave on the basement layer

from the within motion observed at the base of an embankment dam was proposed. The procedure was

apphed to the estimation of the incident wave at a

rockfill dam site during the Kobe Earthquake.

KEYWORDS: Dam, FEM, Earthquake Record,

Dynamic Interaction, Incident Wave

1. INTRODUCnON

Earthquake observations at the bedrock of dam sites are precious data of earthquake motions at the hard

rock site condition, for wliich the S-wave velocity is over about 800 m/s. There are few earthquake

observational stations situated at hard rock sites in Japan, other than dams and nuclear power plants. In most dams in Japan, in particular, seismogr^hs have

been installed in the inspection gallery at the base of

embankment dams or at the bottom of concrete dams

for the purpose of safety control against earthquakes.

Most dams in Japan, except for small earth dams, are constructed directly on the excavated bedrock of

riverbeds and abutments. The lower part of a massive

dam is, roughly considered to behave in the same way as its bedrock and rock abutment during

earthquakes. However, the earthquake response of a dam is affected by the dynamic interaction of the dam-foundation-reservoir system. The earthquake waves observed at the dam base or at the dam bottom

are interaaed especially with its foundation. In other words, the earthquake waves at die dam base are the combination of the incident waves through the foundation and the radiation waves through the dam

body. They are, therefore, different from the waves at the free-field, i.e. outaop. In order to make practical use of the observations at the dam base of the many

dams where seismometers have already been

installed, it is necessary to evaluate the quantitative difference of the earthquake wave at the dam base

with that at the free-field.

The effects of the dynamic interaction between a dam

and its foundation are evaluated by wave energy

dissipation through the dam to the foundation or by

radiation damping. Analytical approaches of

radiation damping of embankment dams have been

perfonmed (Chopra & Perumalswami 1969,

Ohmachi 1980, Yanagisawa 1982, Hirata 1989,

Tohei & Ohmachi 1990). In these evaluations,

analytiical procedures were mainly used and few

comparative evaluations with actual dam

observations were conducted.

In this paper, we evaluate the effects of frequency

contents of input waves and the impedance ratio on

1) Head, Fill Dam Division, Public Works

Research Institute, Ministry of Construction,

Tsukuba, Japan

2) Senior Research Engineer, ditto

253
the dynamic interaction between an embankment

dam and its foundation. The httie effect of a reservoir on an embankment is not considered. The

evaluations are performed using botii FEM numerical

analysis and the earthquake observations at an

existing dam. Moreover, we propose a simplified

analytical procedure that estimates tiie incident seismic wave on the basement layer without the dynamic interaction effects from tiie observations at

tiie base of an embankment dam.

2. DYNAMIC INTERACTION AT BASE OF

AN EMBANKMENT DAM

(1) Analytical Method and Cases

We analyzed the two-dimensional FE model of die Miho Dam in Japan, a 95.0 m-high rockfill dam with

a central core, shown in Figure 1. The physical and

dynamic properties of the dam were determined from

in-situ tests and laboratory tests using tiie fill materials of tiiis dam. We used the complex dynamic

response analysis program DENAS, which employs

equivalent linear dynamic soil modeling.

We performed analyses on the five cases shown in Table 1. Where, p d is tiie density of the dam body

of 23 t/m^ , |0 F is tiie density of the foundation of

2_575 t/m^ , Vsd is tiie S-wave velocity for tiie dam

body of 600ni/s , Vsp is tiie S-wave velocity for tiie foundation. In Cases lA and IB, the effect of the frequency contents of the input wave on the dynamic

interaction was evaluated. In Case lA, the acceleration wave which has a maximum value of

138.2gal, is shown in Figure 2, and was observed at

tiie downstream toe of the Miho Dam during the event of March 6, 1996 (JMA magnitude of 5.8) was

used as input. In Case IB, the same wave as in Case

lA, with the time axis stretched by a factor of two

(that is, with twice calculation period) was used as

input. The predominant frequency of the input wave

of Case IB was consistent witii the natural frequency

of tiie dam body. In Cases IB, 2-^, the effects of tiie impedance ratio between the dam body and its foundation were evaluated by changing tiie shear

stiffness of tiie foundation.

Table 1 Analytical cases

S-wave Velocity Impedance Calculation

Case Name for Foundation Ratio Period Items Investigated

VsF(m/s) a At(s)

CASEIA

CASE IB

1950

1950

0.29

029

0.01 0.02

- Frequency contents of input waves CASE 2

CASES

1380

975

0.41 0.57

0.02

0.02

PD'VSD

>- Impedance ratio a = —

PF'VSF

CASE 4 690 0.81 0.02

Core

Fig. 1 Finite element model of the Miho Dam

254
Max:-138.2gal

I.I.I I.I. 0 2 4 6 8 10 12 Time (s)

Fig.2(a) Input acceleration-time history for Case lA

Frequency (Hz)

Fig.3 Amplification spectra for Case lA

(2) Results of the Analysis

a) Effects of frequency contents of input waves The amplification spectra of the response waves at the crest and at the center of the dam base against the input wave for Case lA are shown in Figure 3. The

amplification spectrum for the dam base dips at a

slightly higher frequency than the natural frequency

that is indicated as the predominant frequency of the amplification spectrum at the crest The amplification

spectnmi for the dam base in Figure 3 shows that the seismic wave at the dam base reduces the seismic

power at the frequency range of the natural frequencies of the dam considerably.

Figure 4 shows the Fourier amplitude spectra of the acceleration at the center of the dam base and at the

free-field for Case LA The amplitude spectrum for the dam base is smaller than that for the free-field for the second and third natural frequencies of the dam

(around 3 Hz). This is because the spectrum of the input wave has large amphtudes around the second

and third natural frequencies of the dam. The Fourier

0 2 4 6 8 10 Frequency (Hz) Fig2(b) Fourier amplitude spectrum of input

acceleration for Case lA

30|

0 2 4 6 8 10 Frequency (Hz)

Fig.4 Fourier ampUtude spectra of acceleration

at dam base and at free-field(Case 1A)

0 2 4 6 8 10

Frequency (Hz)

Fig.5 Fourier amplitude spectra of acceleration

at dam base and at free-field(Case IB)

amplitude spectra for Case IB are shown in Figure 5. TTie amplitude spectrum for the dam base is considerably smaller than that for the free-field around the first natural frequency of 1.7Hz. When an

incident wave that has large energy power around the

natural frequency of a dam is input, djoiamic

interaction between the dam and its foundation

makes the spectral amphtude of the response

acceleration at the dam base decrease around this frequency.

255
Frequency (Hz)

Fig.6 Amplification spectrum ratio for dam base

b) Effects of impedance ratio The amplification spectrum ratio Ar was calculated by dividing the amphfication spectaim for the dam

base by the one for the free-field. The amplification

spectrum ratio Ar shows the difference between the seismic response at the dam base and that at the

free-field. Figure 6 shows the amplification spectrum

ratios for the central part of the dam base for Cases

IB, 2 to 4. Figure 7 shows the amplification spectrum

for the crest. Figure 8 shows the relationship between

the frequency where the amplification spectrum ratio for the dam base sinks in Figure 6 and the predominant frequency of the amplification spectrum

for the crest in Figure 7. Figure 8 shows that each

sinking frequency for the dam base is sHghdy higher

than the corresponding predominant frequency for the crest. Figure 6 shows that the larger the impedance ratio, the larger the dip in the amplification spectrum ratio. In other words, the

larger the impedance ratio, the larger the effect of

dynamic interaction. Figure 7 shows that as the impedance ratio decreases, so the response

magnification of the dam increases. This indicates

that the smaller the impedance ratio, the larger the

non-linearity of the dam body and the lower its natural frequency. It is, therefore, suggested that the sinking frequency for each amplification spectrum

ratio shown in Figure 6 becomes higher as the impedance ratio increases.

Figure 9 shows the distribution of maximum

acceleration, which is normalized by dividing by the maximum acceleration at a free-field, along the dam

base and on the downstream ground surface for each

of the analytical cases. The maximum accelerations are smallest at the center of the dam base and recover

10 CASE1B CASE3 CASE2_ CASE4

0 —' ' < 0 2 4 6 8 10 Frequency (Hz)

Fig.7 Amplification spectrum for crest

as you move toward the downstream toe. The

maximum accelerations at the downsdream toe are only about ten percent less than the maximum

accelerations at the free-field. The larger the impedance ratio, the more the drop of the maximum

accelerations at the dam base becomes and the larger the dynamic interaction between the dam and the foundation becomes.

The residual of the amplification spectrum ratio was

integrated from 0.5fl to 2.5fl, where fl is the first natural frequency. The mean residual e shown in equation (1) represents the ratio of the energy loss at

the dam base due to the dynamic interaction.

2fijoofi

Figure 10 shows the distribution of the mean residual e along the dam base (footprint) and on the downstream ground surface for each of the analytical

cases. The mean residual e is large around the center of the dam base and becomes smaller as you move

toward the downstream toe. Furthermore, the larger

the impedance ratio, the larger the mean residual, i.e.

the larger the energy loss due to dynamic interaction. Hie values of the mean residual e are only about 0.1 around the downstream toe for all the cases.

Figures 9 and 10 indicate that the effect of dynamic

interaction on earthquake response is low around the downstream toe, regardless of the impedance ratio. There is negligible effect of dynamic interaction due

to the dam on the seismic response at the point about

half a dam's width away from the downstream toe.

256
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1996

CASE1

A

CASEIB

CASE2

CASE3

CASE4

Observations e 12 3 4

Predominant Frequency of Amplification

Spectmm at Crest (Hz) Fig.8 Comparison between sinking frequency of

tansfer function ratio for dam base and predominant

frequency of transfer function for crest

(3) Verification of Analytical Results Using

Observations

The characteristics of the earthquake motions at the dam base obtained from the analytical procedures in part 2.(2) above were verified using the earthquake

observations duuing three events shown in Table 2

and the data of microtremor measurements at the Miho Dam. We calculated the ratio of the Fourier

amplitude spectrum of the waves observed at the

crest to that at the outaop. We also calculated the

ratio of the Fourier amphtude spectrum of the waves

observed at the center of the dam base to that at the

outCTop. The relationships between the sinking

frequency of the Fourier amphtude spectrum ratio for the center of the dam base and the predominant

Distance from the Dam Axis, L

Fig.9 Distribution of maximum acceleration at dam

base and downstream ground surface

Distance from the Dam Axis, L

Fig. 10 Distribution of mean residual of amplification

spectrum ratio for dam base and downstream ground

surface

frequency for the crest obtained from the observations and the microtremor data have been

added to Figure 8. The points appear shghtly above

the 45-degree line that represents the case where the sinking frequency for the dam base is equal to the predominant frequency for the crest. In other words,

the sinking frequency for the dam base is shghtly

higher than the corresponding predominant

frequency for the crest, which agrees well with the

analytical results in part 2.(2) above.

Table 2 Earthquake records at the Miho Dam

Date Maximum Acceleration (gal)

Crest Gallary Downstream Toe

Epicentia] Distance

(km)

JMA Magnitude

March 6,1996 243.9 84.4 1382 13 5.8

August 9, 19% 623 13.8 21.8 17 4.7

October 26,1996 15.9 3.4 5.4 11 4.0

257
The normalized maximum acceleration and the mean

residual of the amplification spectrum ratio from the earthquake observations during the event of March 6, 1996 and the microtremor data measured at the center of the dam base and the downstream toe have

been added to Figures 9 and 10 respectively. The

waves observed at the outcrop were used as the reference wave records. The maximum acceleration

at the center of the dam base is smaller than that at the downstream toe. The mean residual e at the center of the dam base is larger than that at the downstream

toe. These trends agree with those seen in the

analytical results.

3. ESTIMATION OF SEISMIC WAVE

INCIDENT ON BASEMENT LAYER

(1) Estimation Procedure

As stated in the previous section, earthquake waves

observed at the dam base are interacted with the dam

body. Hence, we propose a simplified procedure for estimating the wave incident on the basement layer

from observations at the base of an embankment dam

analytically. Ohmachi & Kataoka (1995) estimated

the incident seismic waves from the observations at the base of a concrete dam using the procedure that involved dividing the frequency characteristics of the observations at the dam base by the transfer function. However, an earth structure such as an embankment

dam has the high earthquake-induced non-linearity of

the soil and rockfill materials. This is the major

difference between an embankment dam and a conaete dam with regard to the estimation of the incident wave for earth structure. For this reason, we

performed dynamic equivalent linear pre-analysis on

the model of an embankment dam with a rigid

foundation. A rigid foundation model is analyzed

because the wave observed at the dam base must be

input. This dynamic pre-analysis gives the shear modulus of the dam reached in the inelastic region

during shaking. Furthermore, a comparison between

the analytical response results and other observations

at the dam enables the accuracy of the dam model to be verified.

The estimation procedure is illustrated in Figure 11 and described as follows:

STEP 1 . The dynamic pre-analysis of the embankment dam model with a rigid foundation is performed using the equivalent linear soil modeling

method. The distribution of the shear modulus of the dam in the final iterative calculation step is obtained. STEP 2. The transfer function for the dam base is calculated using the above shear modulus of the dam.

STEP 3 . The Fourier spectrum, which involves the

characteristics of both amphtude and phase, of the wave incident on the basement layer is obtained by

dividing the Fourier spectrum of the wave observed

at the dam base by the above transfer function. STEP 4. The Fourier inverse transform of the Fourier

spectrum calculated above gives the seismic wave

incident on the basement layer.

An observed wave consists of the coupled response

of horizontal and vertical vibrations. In this simphfied

procedure, however, we consider only the horizontal

transfer function for the horizontal input motion and

only the vertical transfer function for the vertical input motion.

(2) Dam Analyzed and its Model

We apphed this procedure to the earthquake

observations at the Minoogawa Dam during the Kobe Earthquake of 1995. The Minoogawa Dam, a 47.0 m-high rockfill dam, was located about ten kilometers northeast of the earthquake fault. The

seismometers installed in the inspection gallery at the base of the dam and at the dam crest recorded

acceleration waves during the excitation. Their

maximum horizontal accelerations perpendicular to the dam axis were 135 gal in the gallery and 242 gal

at the cresL The dam foundation is hard bedrock that has an elastic wave velocity of about 4 km/s. The

impedance ratio between the dam and the foundation

is about 0.3.

258
Pre-analysis sTEPl

Shear Modulus

Observed Wave Fourier Spectrum of Observed Wave at Dam Base

(Involving characteristics of \

botti amplitude and phase )

-Oi

Dam Base (Footprint)

Compliant Foundation

Amplification Spectrun^ Dam Base

Basement Layer

Basement Layer)

Deconvolution

>tep3

Incident Wave

step4

Fourier Spectrum of Incident Wave on Basement Layer

Fourier Inverse

Transform

-co

Fig. 1 1 Illustration of the procedure to estimate incident waves

We prepared the 2-D FE model of the cross section of the Minoogawa Dam as shown in Figure 12. The

properties of the dam materials were determined

from the results of execution control tests in the time

of the construction and the laboratory tests using the

fill materials of this dam. We verified the accuracy of

the vibration characteristics of the model by

comparing the natural frequency of the model with

that from earthquake observations and microtremor measurements (Iwashita & Yoshida 1997). In the pre-analysis of the rigid foundation model, the effects of energy dissipation by dynamic interaction were

assumed to be represented by the equivalent radiation damping ratio that was added to the hysteresis

damping ratio of the dam materials.

Core

(3) Calculation of an Incident Wave

First, we analyzed the rigid foundation model and

obtained the shear modulus of the dam during

excitation. Figure 13 shows comparisons of the observed response acceleration at the crest with that obtained through the dynamic pre-analysis. Lxwldng

at the acceleration-time history, during the first half of

the principal motion (t = 4 - 6 sec) the analytical

results agree weU with the observations, but during

the second half (over 6 sec) the analytical results are smaller than the observed values. With regard to the Fourier ampUtude spectrum, the analytical results in the low frequency range less than 2 Hz agree well with the observations, but the analytical results are

Filter

Energy TVansmittiiig Boundary

/

Fig. 12 Finite element model of the Minogawa Dam

259
300

200

100

0

-100

-200

-300

Observation

Analysis

Perpendicular to Dam Axis Observation

Analysis

Perpendicular to Dam Axis

Frequency (Hz)

Fig. 13 Comparison between observed acceleration at crest with analytical results (pre-analysis of the Minogawa Dam model with a rigid foundation)

6 8 10 12 14 16 18 20 Frequency (Hz)

6 8 10 12 14 16 18 20 Frequency (Hz)

Fig. 14 Amplification spectra for the center of the dam base of the Minogawa Dam model

smaller than the observed values around the natural frequency during shaking of 3 Hz. In spite of the

rigid foundation model analysis, the analytical results give relatively good agreement with the observations. This confirms that the model describes the

non-linearity of the dam during the earthquake quite

well.

Secondly, the transfer functions for the dam base were calculated using the shear modulus of the dam,

which is the output fi-om the above pre-analysis on

the rigid foundation model. The ampUfication spectra

shown in Figure 14 have significant sinking at 3 Hz

for the direction perpendicular to the dam axis and at about 4Hz for the vertical direction.

Finally, we estimated the wave incident on the basement layer of the Muioogawa Dam site using

steps 3 and 4 of the above-mentioned procedure, with

the earthquake observations in the inspection gallery

built on the dam base and its transfer functions.

Figure 15 shows that the estimated incident wave

time history and Fourier amphtude spectrum. The

wave observed at the inspection gaUery is also shown

in the figures for the shake of comparison. The

horizontal maximum accelerations of the incident wave and the wave observed in the gallery are 137

gal and 135 gal respectively; the vertical maximum

accelerations are 89 gal and 80 gal respectively. The

amphtude spectrum of the incident wave is larger than that for the wave observed at the gallery at the

natural firequencies (the range around 3 Hz for the horizontal direction and 4 to 5 Hz for the vertical

direction) because of the dynamic interaction effects described in section 2.

(4) Verification of the Estimated Incident Wave

We performed dynamic analysis on the Minoogawa

Dam model with a comphance foundation. The

incident wave estimated in the previous section was

input under the basement boundary of the

260
Estimated Incideni wave on basement

Otssefvabon at center

ot dam tiase

Perpendicular to Dam Axis

2 4 6 Time (s)

Perpendicular to Dam Axis

Estimated indtJent wave on basement

Observation at center

of dam base

4 6

Frequency (Hz)

Estimated mcideni wave on basement

Otjsen/ation at center

of dam base 1^

Vertical

1 . t .

Vertical

Estimated inadent wave on basement

Obsen^tiOn at center

of dam base

Time (s) Frequency (Hz)

Fig. 15 Estimated incident acceleration on basement layer and observed acceleration at dam base

400

200

-200

-400

Anaiysis

Observation

'erpendic

III

ular to Dam Axis

11 f ' '

4 6

Tiroe (s)

10 4 6

Frequency (Hz)

Fig. 16 Corr^arison of observed acceleration at crest with analytical results (analysis on the Minogawa

Dam model with a compliant foundation ,on which the estimated incident wave was input)

compliance foundation model. Figure 16 shows the comparisons of the acceleration-time histories and

Fourier amplitude spectra of the analytiail response

horizontal acceleration at the a-est with those of the observed acceleration at the crest. The results of the

analysis agree with the observations for both the time

history and the Fourier spectrum. Of particular note is the good agreement of the time histories even for the second half of the principal motipn and of the Fourier

amplitudes at around 3 Hz. In the above range, the

results from the simulation of the rigid foundation

model analysis did not agree with the observations, as shown in Figurel3.

4. CONCLUSIONS

(1) Due to the dynamic interaction between an embankment dam and its foundation, the Fourier

amplitude of the seismic waves at the dam base

at the natural frequency is smaller than that at the

free-field. It was demonstrated from FEM

analysis and observations at an existing dam Uiat the sinking frequency is slightly higher than the conresponding predominant frequencies for the response wave at the crest.

(2) The larger the Fourier amphtude of the incident seismic wave for the natural frequency of the

261
dam and the larger the impedance ratio of the dam and its foundation, the larger the dynamic

interaction effects become.

(3) Dynamic interaction effects such as reductions in maximum acceleration and wave energy are the

largest at the center of the dam base and decrease

as you move towards the downstream toe. These

interaction effects are fairly small by time you

reach the toe. The effects at a point about half a dam's width firom the dowTistream toe are about

the same as those at a free-field.

(4) A simplified procedure for estimating the

incident wave from the observations at the dam

base was proposed. This procedure was

demonstrated by estimating the waves incident on the basement layer at the Minoogawa Dam

site during the Kobe Earthquake.

REFERENCES

Chopra, A.K. & P.R. Perumalswami (1969):

Dam-foundation interaction during earthquakes,

Proc 4th World Conference on Earthquake

Engineering, pp.37-52.

Hirata, K. (1989): Estimation of radiation damping in dynamic analysis of fill dams. Report of Central

Research Institute of Electric Power Industry.

U88061, CRIEPI.

Iwashita, T. & H. Yoshida (1997): Earthquake

response analysis and seismic stabihty of a rockfill dam -Behavior of the Minoogawa Dam during the Kobe Earthquake-, Engineeringfor Dams. No.l26,

pp.27-35, Japan Dam Engineering Center. Ohmachi, T. (1980): A fundamental study on

dynamic interactions with a rockfill dam and the foundation deposit, Tsuchi-to-Kiso, No.ll91,

pp.31-36, Japanese Society odf Soil Mechanics and

Foundation Engineering.

Ohmachi, T. & S. Kataoka (1995): Evaluation of

dynamic interaction effe-cts of 2-D

dam-foundation-reservoir systems, fVoc. of the Japan Society of Civil Engineers, No-519:

pp. 199-209, Japan Society of Civil Engineers.

Tohei, M. & T. Ohmachi (1990): Model analysis

procedure using a FE-BE method in time domain

and its application to a dam-foundation system,

iVoc. of the Japan Society of Civil Engineers, 416,

pp.429-438, Japan Society of Civil Engineers.

Yanagisawa, E. (1982): Effect of ground condition on vibrational characteristics of earth structure, Proc. of the Japan Society of Civil Engineers,

No.317, pp.101-110, Japan Society of Civil Engineers.

262
REPRODUCTION OF LARGE-SCALE IG TEST ON DRY SAND DEPOSITS AND PILE

FOUNDATIONS USING CENTRIFUGE MODELING (Step-2)

Chikahiro MINOWA", Masayoshi SATO", Yukiko SAITO'', Takaaki KAGAWA^' and Akio ABE''

ABSTRACT

The two centrifuge tests, the SC-centrifuge test and the PW-centrifuge test, were executed to reproduce the dynamic behavior of pile foundation structures and dry sand

deposits in an NIED-lg large-scale test. The following

conclusions may be obtained from a comparison of the

three tests, (1) the PW-centrifuge test showed a good cor- respondence with the NIED-lg test regard to the input acceleration wave, (2) the PW-centrifuge can virtually repro- duce the NIED-lg test m terms of the structure response

acceleration and the bending strains of the pile, (3) centri- fuge studies could confirm the composition of similitude requirements and the effectiveness as a preliminary study

method for large-scale tests.

KEY WORDS ; Large-scale shaking test. Centrifuge dynamic testing, Dry sand,

Ground response, Soil-pile interaction

1. INTRODUCTION

Dynamic centrifuge model test technology has recently

attracted considerable attention in the field of geotechnical

earthquake engineering, mainly because it has the obvious

advantage in being able to reproduce the dynamic response

of actual soil ground under the conditions of satisfying the

similitude requirements concerning confined stress of soil. However, when soil models are prepared for centrifuge tests,

they may not always be perfectly consistent with the actual

soil ground. It is important to establish test procedures and

techniques for reproducing the behavior of actual soil-structure systems during earthquakes. Recently, there are many

plans to execute large-scale shaking table tests. We cannot perform the large-scale tests for many cases, since they

generally require considerable funds and long periods to

execute. Among the numerous potential case scenarios, it

is very difficult to decide which cases are important. Accordingly, there is considerable validity in selecting rea- sonable experimental cases of the large-scale test from the execution of centrifuge tests with many parameters. The

purposes of the centrifuge studies are to check similitude requirements and to establish a preliminary study method

for large-scale tests.

In the past, a dynamic centrifuge test using Shimizu Corporation centrifuge apparatus (SC-centrifuge) satisfying

the similitude requirements was conducted in reference to a large-scale test with laminar containers. The centrifuge

test could not reproduce the response of the large-scale

test result, therefore could not reproduce the input wave.

In this paper, a test was executed using Public Works Re- search Institute centrifuge apparatus (PW-centrifuge) that are capable of sufficiently reproducing mput waves, so the comparison of the three tests results, NIED-lg large-scale

test, SC and PWRI centrifuge tests, are reported in respect

to the dynamic behavior of pile foundation structures and

dry sand deposits.

2. SHAKING TABLE TEST APPARATUS

(1) Test apparatus

A large-scale shaking table test using a large laminar conTable 1 Specifications of the three test apparatus

Unit NIED-lg

SCcentrifuge PWcentrifuge

Table size m 15x14. 5 0. 65x0. 95 1. 0x1. 7 Maximum loading capacity

t 500 0. 3 4

Excitation

direction

1 1 1 Maximum

acceleration

gal 500 9. 800 39. 200 Maximum

displacement mm 230 ±1 ±5

Frequency range Hz DC~50 50~350 10~400

Radius of rotation mm 3. 1 6. 6

Centrifuge

acceleration g 50 100

1) National Research Institute for Earth Science and Disaster Pre- vention, 3-1 Tennoudai, Tsukuba-shi, .105-0006, Japan.

2) Public Works Research Institute, Ministry of Construction,

Tsukuba-shi, 305-0804, Japan.

3) Wayne State University, Michigan, 48202, USA.

4) Tokyo Soil Research Co., Tsukuba-shi, 305-0045, Japan.

263
Fig3 Laminar container of centrifuge test at Shimizu Corporation

264
Sbiking direciion

; Aocckromcicrs

: DispUocmeni transducers

; Strftin gages

Shakmg 7Ale

Elan.

Sitcl pilf-- 0 = 318.5. 1=6.

IXi-Hl

2.io

11,600

CIC2

G304

C5C6C7

GSOTcm.

A'Sl

1 c-1- C- 14 • i

6,000

3.100

^fr! inn

Displacenoni iransduccrs

I Si rain gages

JA-SI

Dis-Hr

1 A-SI

I I*

6 = 20, 1=0 5

-G-7

SOO

Fig. 4 A large-scale shaking test model and locations of transducers

-100| ^^o-^

80 - ^0 =0. 27BIII Uc = 3 60 - Cs = 2 72

Fc = 3 OX

«>c 40

= 1 74

= 1 39 20

0 001 0. 01 0. 1 1 10 Grain size (mm)

Fig. 5 Grain-size accumulation curve and physical

properties

tainer was conducted at the National Research Institute for

Earth Science and Disaster Prevention (NIED). The table is

15.0 m X 14J m in size and its maximum payload is 500 1. The first centrifuge model tests were performed at the Institute of Technology, Shimizu Corporation (SC) [1]. The

specifications have been described in detail by Sato [2). Additional dynamic centrifuge model tests were performed

at the Public Works Research Institute (PWRI). The PWRl

centrifuge apparatus is shown in Fig. 1. TTie specifications

have been described in detail by Matsuo et al. [3]. The

specifications of the three test apparatuses are shown in Table 1.

Fig. 6 A the centrifuge shaking test model and locations

of transducers

(2) Large-scale laminar container

The NIED-lg large-scale laminar container is shown in

Fig. 2. The container has inner dimensions of 11 .6 m L x 6.0 m H X 3.1 m W. The container consists of 29 rectangular frames made of H-shaped steel members (200 mm x 200 mm).

(3) Centrifuge laminar container

The same laminar container was employed for both the

SC-centrifuge test and the PW-centrifuge test. The centri- fuge laminar container is shown in Fig. 3. The laminar

container, which is geometrically similar to the one used in

the large-scale shaking test, has inner dimensions of 80 cm Lx41cmHx23cmW.

3. LARGE-SCALE SHAKING TEST

(1) Preparation ground

A NIED-lg large-scale test model and the locations of

transducers are shown in Fig. 4. The test sand was poured

through a funnel from a height of 1 m. The sand deposit had an average density of P t=1.69t/m3 and a water content of w = 1 1 % at the beginning of the test. The grain-size

265
Table 2 Similitude requirements used in the experiment

Svni t>ol Sole

ntio U nit Centri f ugf

Sand Thickness H

f 1/^ m 5 981 0J99

Sirs Cum

P, 1 g/cm* 1.69 1.69

Length of pile L MX m 60 0.4

Diamcicr D MX mm 318-5 20

(21.4)

Thickness I MX rrm 6.9 OS

(0.46)

RIe Young's modulus E 1 MN/nf 206.000 206.000

Gcorrrirical mDmcni ofincrlia

1 1/A ' Ctrl' 8.200 0 162

Bending sJiflncss E 1 -[fx' MNrrf 16.9 0.000300

(0.000334)

Area A MX' ctn- 67^5 0300

Normal sii&icss EA

MX' MN 1 J92 6.18

Structure Mass MX = 15,600 4.62

Lcnglh L, MX tn 2S 0.167

Exciting acceleratJon o X g

0.6 9 0 nModel/Prolotype=l/ A=l/15

accumulation curve and physical properties of the materials

are sho\ATi in Fig. 5.

(2) Pile foundation structure The founciation is a rigid structure, consisting of slacked

rectangular steel plates, with a length of 3.0 m, a width of 2.5 m, a height of 0.26 m, and weighting 15.6 1. Four (2x2) actual

steel piles with an outer diameter of 318.5 mm were used

with a center-to-cenler spacing 6 times the diameter. The

top of each pile was firmly fixed to the structure, and the bottom of each pile was pin-connected to the container base.

(3) Excitation

Excitation cases of the NIED-lg large-scale tests are described in reference [4]. In this paper, the test result of the

conditions is reported: a structure weight of 15.6 t, a sinusoidal wave with incremental acceleration, an excitation fre- quency of 3 Hz, and a maximum acceleration of 600 gal.

4. CENTRIFUGE SHAKING TABLE TESTS

(1) Similitude requirements

The similitude requirement used in the centrifuge test is summarized in Table 2. The ratio of the centrifuge test and

the large-scale test is 1/15. In the time history and strain diagrams shown in this paper, the values of the centrifuge

test were translated to actual size based on the similitude

requirements.

(2) Preparation of model ground

The soil density, the natural frequency of the ground, the

specifications of the pile foundation, and the input wave of

the large-scale test were known, as necessary data for the execution of centrifuge experiment reproducing the largescale test results were obtained in advance. The centrifuge shaking test model and locations of trans- ducers are shown in Fig. 6. The same sand as that used in

the large-scale test was utilized , and a model with the same

preparation method employed. The natural frequency of

the model ground in the centrifuge test was made equal to

that in the large-scale test.

(3) Pile foundation structure The same model of the pile foundation structure was

employed in both the SC-centrifuge and the PW-centrifuge

test. Model iron piles satisfying the similitude requirements

in respect to bending stiffoess were adopted. Each pile has

a diameter of 20 mm, a wall thickness of 0-5 mm, and a length

qf 40 cm. The same conditions as those in the large-scale

test were adopted regarding the number of piles, pile ar- rangement, and the fixed method of the pile top and bottom.

(4) Excitation The input wave was reduced to the time scale of that used in the large-scale test to 1/15 respectively in the SCcentrifuge and the PW-centrifuge test.

266
too

< 0.1

0.01 1 —' :

:

;

1—

I 5 10 >

Frequency (Hz)

(a) SC-centrifuge test

100

0.01 I

'—' 1——

1 5 10 50

Frequency (Hz)

(c) the NIED-lg large-scale test

Fig. 7 Comparison of the frequency transfer function for the large-scale shaking test and the two centrifuge tests

5. COMPARISON OF EXPERIMEmAL RESULTS

(1) Natural fi-equency of sand deposits

MicTOlremors were measured to investigate the characteristics of the sand deposit before the start of the largescale test, and the natural frequency was found to be 5 Hz

as shown in Fig. 7 (c). Due to the fact that microtremors could not be measured

in the centrifuge tests, a vibration test with a very small

acceleration level of excitation was conducted. The natural frequencies of the three test results are in good agree- ment. Accordingly, it can be estimated that the three sand

deposits have almost identified dynamic response characteristics in terms of the initial shear modulus.

(2) Comparison of input waves The time histories of the input waves for the NIED-lg

600

Fig. 8 Comparison of time history of the input

acceleration for the large-scale shaking test and the two centrifuge tests

1000, ^ , ^ . ,

. . ^

Fig. 9 Comparison of the time history of the structure

acceleration for the large-scale shaking test (A-Sl) and the two centrifuge tests (A-Sl)

large-scale test and the two centrifuge tests are shown in

Fig. 8. The input acceleration in the SC-centrifuge test is lower than the NIED-lg test, but the PW-centrifuge test shows good agreement.

(3) Acceleration responses of the pile foundations and the sand deposits

The comparison of the time history of the three test results in terms of the structure acceleration (A-Sl) is shown

in Fig. 9. Both response accelerations of the SC-centrifuge

and the PW-centrifuge are lower than those for the NIED267
<300

Fig. 10 Comparison of the time history of the soil surface

acceleration for the large-scale shaking test (A-Gl) and the two centrifuge tests (A-Gl)

600 ^300

<300

-600 600 ^300

0 300

-600 600

^ 0 ^00

-600

SC-Cen t r i f uge tes f

i

PW-Cent ri fuge test vwW/Wj

0 5 -r-

, . 10 Time (s)

' . . 1

1 . ,

NIED-lg test

15

Fig. 11 Comparison of the time history of the soil midJe- part acceleration for the large-scale shaking test (A-G4)

and the two centrifuge tests (A-G3)

„ 700

e 0 ^-350

CO -700 _ 700

'° 350

r-350 ^-700 _ 700 o 350

0 2-350 ^-700

SC-Cen t r I f uge test ^vwwWWW

PW-Centri fuge test vAM'V\

NIED-lg test

0 5 10 Time (s) 15

Fig. 12 Comparison of the time history of the bending

strain of piles for the large-scale shaking test (G-1)

and the two centrifuge tests (G-1)

_ 600 2 300 1 0 ^-300 "-600 „ 600 2 300

"

0 c -300 '"-600 _ 600

300 ^ 0

c"-300 ^-600

SC-Cen t r i fuge tes t

—1 1 . . 1 r- PW-Centri fuge test ^m^m!\MI^

- NIED-lg test

0 5 ^. , . 10 Time (s)

-T 1 1—

15

Fig. 13 Comparison of the time history of the bending

strain of piles for the large-scale shaking test (G-10) and the two centrifuge tests (G-3)

Ig test at the top of structure. But the test result of the PWcentrifuge corresponds to that of SC-centrifuge.

The comparison of the time history of the three test results in terms of the ground surface acceleration (A-Gl) is shown in Fig. 10 and for ground acceleration in Fig.l 1. The

three tests resulted in similar soil surface response accelerations. The disturbed shape of response acceleration wave after 9 seconds can be interpreted as degradation of

soil stiffness in the intermediate depth of the NIED-lg test.

In the next study, we intend to confirm the phenomenon

using the nonlinear response analysis method.

(4) Beading strain of piles

The comparison of the time history of the three test results in terms of the pile top bending strains (G-1) is shown

in Fig. 12 and that in the middle depth in Fig. 13. The distri- butions of bending strain for the three tests when large values were observed for the piles are shown in Fig. 14.

Incidentally, the relationship of bending strain and bending

moment is desaibed in reference [1]. The bending strain in the PW-centrifuge test at the top of the pile and the intermediate height were slightly larger than those in the NIED-lg test in terms of the time histories

268
0

« 1 —•—

1 - Ig test —A

- SC-Centr ( fuge - Pl-Centr i fuge

1 1 -800 -400 0 400 800 Bending strain (xlT*)

Fig. 14 Comparison of the distributioa of the bending strain of piles for the large-scale shaking test and the wo centrifuge tests

and distribution, and the SC-centrifuge test results were

smaller than the NIED-lg test overall. It can be concluded

that the PW-centrifuge test virtually reproduced the NIEDlg large-scale test.

6. CONCLUSION

The two centrifuge tests, the SC-centrifuge test and the PW-centrifuge test were executed to reproduce the dynamic behavior of pile foundation structures and dry sand

deposits in an NIED-lg large-scale test. The following

conclusions may be obtained from a comparison of the three

tests.

(1) The PW-centrifuge test was in good agreement with

the NIED-lg test in terms of the input acceleration wave.

(2) The PW-centrifuge test can virtually reproduce the NIED-lg test in terms of the response acceleration of the

structure and the bending strains of the pile.

(3) Centrifuge studies could confirm the composition of

similitude requirement and the effectiveness as a prelimi- nary study method for large-scale tests.

dynamic geotechnical centrifuge at PWRl", Proceedings of The

Internalional Conference Centrifuge 98, Tokyo, pp. 25-30, Baikema.

4) Naganawa, T, Y. Muruno, H. Wang, A. Nishimura, and C. Minowa,

(1998) : "L^rge-scaie Shaking Table Tesis for Foundations and

Buried Structures (Part-7) Vibration Tests on Steel Pile Foundations in Sand Deposits", Proceedings of 33th Japan National Conference on Geotechnical Engineering, Vol.1, pp. 1059-1060 (in

Japanese).

REFERENCE

1) Sato. M., T. Kagawa and C. Minowa (2000): "Reproduction of

Large-scale Test on Unsaturated Sand Deposit and Pile Foundation using Centrifuge Modeling", Proceedings of The 12th World Conference on Earthquake Engineering, Paper No. 2093 (CDRom), Newziland

2) Sato, M. (1994): "A new dynamic geotechnical centrifuge and

performance of shaking table tests". Proceedings of The International Conference Centnfuge 94, Singapore, pp. 157-162, Baikema.

3) Malsuo, O , T. Tsulsumi K. Kondoh and S. Tamoto (1998) : "The

269

An experimental study for the critical state of Footing

loaded moment and shear by tension

by

Shoichi SAEKI Jiro FUKUI , Masahiro SHIRATO -*

, Shin-ichiro FURUSHO

Abstract

In footings, tensile fiber stress results from

bending caused by tensile forces from

pull-out piles during earthquakes.

However, a reasonable calculation method

has not been established for the bending

and shear capacity of footings in such

condition yet. In this paper, we describe

the result of a loading test using model

footings supported piles to investigate the

failure behavior of the footings whose top

is tense. The loading test consists of the

following four cases; two cases are to investigate bending failure behavior and

the others are to investigate shear one. The experimental result shows that the

footings resist with their almost full width

at and after maximum strength and with a

tied arch mechanism especially for shear even when their reinforcement bar at the

top is the main reinforcement bar.

substantial increase of seismic force

accomplishing the introduction of the

ductility design method.

When a bridge is subjected to horizontal

seismic force, we can imagine two design

situations for the footings of pile

foundations or the top slabs of caisson

foundations. One is tension fiber stress occurs at the bottom of them and another

is at the top of them. In the former case, many loading tests of RC deep beams and

deep slabs were performed in the past and

the failure mechanisms were clarified well

in moment and shear by compressive force.

Based on these studies, the specification

prescribes that the full footing width is effective in resisting moment and shear,

and also the increase of the shear strength

with respect to a RC slender beam

according to shear span ratio by taken the

arch action into account.

Key words- footing, highway bridge,

moment, shear, and ultimate strength

1. Introduction

After the Hyogo-ken Nanbu Earthquake,

the ductility design method was specified

for not only piers introduced already but

also foundations newly in the Specifications for Highway Bridges in Japan. For

the design of footings and top slabs, the

allowable stress design was changed into

the ultimate strength design due to a

On the other hand, in the latter case where

the top becomes tense, we have to design

the footings and the slabs for the tension

forces from the columns and piles.

However the strength for those tension

forces was not usual design factor in the

design of the footings with extremely large

dead load like bridges when the design

1) Public Works Research Center,

1-6- 4 Taito Taito-ku Tokyo Japan

2) Public Works Research Institute,

1 Asahi Tukuba-shi Ibaraki-ken

Japan

271
seismic force was small like before the

HyogQ-ken Nanbu Earthquake, almost no

loading experiments assuming such

condition have been performed and the

failure mechanism in this condition is not

clarified to date. Consequently, Specifications for Highway Bridges recommends

that, in the case of that tensile fiber stress occurs at the top of the footings, the

effective width 'b' in the calculation of

bending strength is still b = tc + d as the

same manner in the allowable design

method. Where 'to' is the column pier width,

'd' is the effective depth of the footing. And

also it requires that the shear strength

should be estimated as a slender beam

assuming the full width to be effective.

This means that even if the design seismic

force increased, the design ultimate

strength does not increase. As a result, the area of cross section and the volume of the

reinforcement bar of the footings tend to enlarge in spite that the footings have

never experienced severe damage in the

past.

In this paper, we will introduce the result

of a loading test using model

column-footings supported piles, to study

the collapse mechanism of the footings

placing the top in tension. The horizontal

force is applied near the top of the column

to model a seismic response. We made four

specimens. Two models were designed in

order to undergo bending failure. The

dimensions and the volume of reinforcement bars are constant in these two

models, however the diameters and the

spacing of the reinforcement bar were

varied. Other two specimens were

designed to collapse by shear. These

specimens have the different shear span,

namely the distance from the front surface

of the column to the pull-out pile is different, but the other dimensions keep

the same. Based on the experimental

result obtained here, we analyze the strain

distribution of the reinforcement bars at

the top of the footings and the state of

cracks on the top surface of the footing at

each stages of horizontal force. And we

carry out a trial design to clarify the

bearing mechanism for the moment and

shear of the footings whose fiber stress

becomes tense at the top.

2. Experimental Specimens

This experiment was performed to confirm

the bending and shear strength of footings,

when the top of the footing of a pile

foundation was in a state of tension. Table

1 shows the experimental cases and Figure

1 shows the dimensions of the specimens.

The specimens are whole models of

column-footing-pile foundation. The scale

of these models is about 1/5 of actual

bridge piers. Case 1-1 and Case 1-2 was

carried out to confirm the bending failure.

As shown in Figure 2 and Figure 3, while

the amount of footing reinforcement bars

was maintained at almost constant, the

spacing and the diameter of the reinforcement bars were varied between Case 1-1

and Case 1-2, to confirm how these

differences influence the stress

transmission mechanism. Case 2 and Case

3 were executed to investigate the shear

failure. The shear span of the tension side

was varied to investigate influence of the

shear span on the shear strength.

The loading apparatus is shown in Figure

4. To support the specimens, acrylic plates

were installed under the footing

corresponding to the each pile respectively.

272
These acrylic plates have almost the same

vertical rigidity as actual piles under the

assumed similar ratio. High strength steel

bars modeling tension piles were anchored

to the footing through the acrylic plates.

In all cases, only the horizontal force was

loaded. No axial force was loaded on the

column in order that the top of the footing

became tense. All experiment was

performed under the static loading

condition. One-way multi-cycle loading

was performed in the Case 1-1 and Case

1-2. The horizontal displacement when a reinforcement of the footing yielded was

defined as 1 6, and the displacement was

increased in steps of 1 d. For the Case 2

and Case 3, one-way monotonic loading

was performed until the maximum load.

3. Results of the Bending failure

Experiments

(l) General Observations for the Test

Figure 5 shows the load displacement

curves at the loading point for Case 1-1

and Case 1-2. Those maximum loads were

almost same. The failure states were

similar, the bending cracking on the top

surface of the footing was the dominant

damage, and the failure type was bending

failure with yielding of the top reinforcement.

Figure 6 is a schematic diagram of the

crack patterns of the footing top surface. In

Case 1-1, a crack first initiated at the base

of the column at a horizontal load level of

78.5 kN. Then, at 157.0 kN, cracks®

occurred on the side of the footing. New

Cracksd) appeared at 215.8 kN and some

reinforcement bars arranged at the top of

footing yielded at the side of the column at

the horizontal loading level of 279 kN, the

horizontal displacement of 4.88 mm, 1 0 , and d/R = 0.244%. And the cracks®

extended at the maximum horizontal load

of 306 kN. Later at 1.3 d, cracks (3) developed and the cracks evolved on the

top and side surfaces. Then cracks® were

observed at 1.6 d. During the loading,

however, no abrupt decrease of the load

was observed, finally, loading continued

until 269.8 kN, 26 d (126.8 mm), and 6/U

- 6.34%. In Case 1-2, similar damage to Case I'l occurred, the horizontal

displacement when the top surface

reinforcement bar yielded was 14.69 mm,

(5/H = 0.734%, and the maximum load at

294.3 kN. Loading continued until 10 d,

147,2 mm, and 5/H = 7.36%. Photograph 1 exhibits the state of the failure in Case 1-1.

Figure 7 shows the strain distribution in

the loading direction at cross-section A-A

in Figure 2.3. The yield strain of the steel

reinforcement bars was approximately

2000 |i in the material test. At 1 5 in Case

1-1, the strains in reinforcement bars far

away from column were larger than that

close to the column. In Case 1-2, the

strains at poison close to the edge of the

footing were smaller than the strains

around the column. But the difference

between the two cases was small and, the

reinforcement bars yielded across the full

width at maximum horizontal load in both

cases. These indicate that there is no effect

of the reinforcement spacing. If the strain

distribution at 1 (5in Case 1-1 is converted

to a rectangular distribution, the effective

width be becomes tc + 1.5d. The

reinforcement bar yields across full width

after 1 8\n both cases. As a result, we can

conclude that the bending is resisted by

the reinforcement in the full width of

273
footing.

(2) Comparison of the Experimental

Results and Calculated Results

The yield bending moments were

calculated for each of the following

effective widths and the results compared

with the experimental results. The

effective widths were set in the 3 ways

shown below (see Figure 8).

® b = tc + d (recommended in the present

design specification)

@ b = tc + 2d

(3) Overall width of the footing

Where

;

b: footing effective width

tc- column width

d: effective footing depth

The actual strength of the concrete and the

reinforcement bars obtained from material

tests were used to calculate the yield

bending moment. Figure 9 compares the

calculated yield moment Mcy with the

experimental result Meu for the moment at

critical footing section at the maximum

horizontal load. The results are almost the

identical value in Case 1-1 and Case 1-2.

Assuming to be the effective width b was tc + d, the calculated value was about 60% of

the experimental result and a certain

safety factor was involved. In case that b =

tc -I- 2d, the calculated result was lower

than the experimental results. But when it was the full width, almost the same results

were obtained. In the specimens used for

this experiment, the footing width was

relatively narrow in transverse direction.

Therefore, in these results, there was

almost no notable difference between the

effective width of b = tc + 2d and the full

width. If the footing width is larger

compared with the column depth than

these experimental specimens, cracks were

expected to initiate from the side of footing

Considering these things, we can expect

that the effective width will also be greater

than b = tc + 2d.

Figure 10 presents a comparison of the

bending moment at critical footing section

for Case 1-1 under the horizontal

displacement level at Id. Here, Mcy

represents the computed yield moment for

each effective width and Mey represents the

experimental value. According to this,

when the effective width b was tc + 2d or the full width, the calculated results were

estimated greater than the experimental

results and it was in danger.

4 Results of Shear Failure Experiments

(1) General Observations for the Test

Figure 11 shows the load - displacement

curves at the loading point in Case 2 and

Case 3. Photograph 2 shows crack patterns

of footing at maximum load time in Case 2. In both cases, the maximum load was

reached before the tensile reinforcement

yielded. Thus it was considered that the

footing failed in shear. It confirms that in

all cases, after maximum load, the load

declined gradually and even after the

strength declined, a certain degree of

deformation performance was maintained.

Figure 12 shows the state of crack in Case

2. There were two kinds of crack on the

side surface of the footing: an arch-shaped

cracks from the top of the footing of the

pull-out pile towards the compressive edge

of the column (D and to the bottom edge of

the footing on the push-out pile (D, and the

274
ultimate failure was caused by the latter

(2). This indicates shear is being carried by

tied arch mechanism from pull out pile to the compressive side of column through

the bottom of footing under the column.

The figure also confirms that the cracks

extended beyond the critical section B-B

and reach the footing of the portion under

the column. This indicated that the shear

span in the case of the top surface tension

is longer than that of the bottom surface

tension.

(2) Effect of the Shear Span Ratio on the

Shear Strength

We will discuss the effect of the shear span

ratio on the shear capacity of footing based on these results and experimental results

in the past.

Figure 13 shows the relation between the

shear span ratio and Sc / Se, where Sc is the shear strength assuming a RC slender

beam. Se is the experimental result of the

maximum shear force at critical footing

section. Here we would like to define the

shear span, the distance from the center of

the pile to the tensile surface of the column.

The experimental data we conducted in the

past using cantilever beams were also

shown in Figure 13. It elucidates that even

in the case that top is loaded tensile stress,

the shear strength rises as the shear span

ratio decreases.

5. Conclusion

The bending failure experiments have

confirmed that, at the maximum strength,

bending is bore with the almost full width

of the footing even if in the tension side.

These experimental results suggest that

we can design under following method. If

we may permit the advance of cracking to a certain degree, design is based on bending

strength, the 'effective width b may be

equal to or greater than tc + 2d. If harmful

damage in footing isn't permitted,

allowable stress verification may be

performed under effective width b = tc + d.

Judging from the shear failure

experiments, when the top side is in

tension, it is possible to adopt the design

method that the shear strength increases

according to the decrease of shear span

ratio, as in the case of bottom side tension.

But we need to define the longer shear

span than that of compressive side based

on observing shear cracks. We conclude

that footing can develop a tied-arch action

from the crack pattern like an arch shape

observed in Figure 12 and the analysis in

Figure 13. And substantial shear force can

be transmitted by compression force

through the arch rib formed from the top of

the footing at the pull-out pile to the

bottom of the column.

We will conduct further experiments to propose a practical design method for

footing shear strength in the case of top

side tension.

275
Table 1

Case case 1-1 case 1-2 case 2 case3

Foundation type Pile foundation (4 pQes)

t allure mode Bending Shear

rooting shape Symmetncal model Asymmetncal model

Shear span a (mm) 375 600

hnective height dlmm; 450

Shear span ratio a/d 0.833 1.333

Footing reinforcement (top) D10ctc75 D13ctcl50 D29ctc75 D29ctc75

Footing remforcement (bottom) D22ctc75 D22ctc75 D22ctc75 D22ctc75

Steel ratio (%) (top) 0.18 0.17 1.62 1.62

Steel ratio (%) (bottom) 0.98 0.98 0.98 0.98

CASEl-1 CASEl-2 CASE2 CASE3

1.23!

s

P

5

-( ^-

s

-< >- -<

5

^« [_13I1_|£2Q

l.?CC

p

r

r

c

1 -< -< >- e

Figure 1. Dimensions of Specimens

276
Loading direction Loading direction

0

^

il . I . . . .1 ...

I F |« I JT MM

^-IPf 013

A'

Figure 2. Case 1-1 Reinforcement

Loading direct ion

700

Figure 3. Case 1-2 Reinforcement

Acrylic plate^ 200^d=70

High strength steel bar

Figure 4. Test Setup

277
case1-1 (15 =4.88mm)

Displacement (mm)

easel -2 (1(5= 14.69mm)

l5 294kN

^

Displacement (mm)

Figure 5. Load—Displacement Curve

-H'3

--f. \-' 7

t 1i

1

T

ui

/ w

Figure 6. Crack Pattern (Case M)

278
case 1 -1

156.8kN -^215.6kN

• 1 (5(279kN) r35(301kN)

I-6(5(300kN) 2(5(288kN)

3 5(286kN)

4(5(295kN)

150 300 450 600 750 900

Overall width (mm)

1050 1200 1350 1500

case1-2 137.3kN ' 225,6kN

1(5(294kN) i 1.5(5(281kN) 2(5(243kN) 3(5(249kN)

-4(5(240kN)

600 750 900

Overall width (mm)

1500

Figure 7. Strain Distribution at A-A' section in Figure 2,3 (Case 1-1 and Case 1-2)

n2501 725 725 |250

8 ^

in + — u 4— O

a + 1,300 +j

o

in

^^ u oos

oo

in

Figure 8. Obtaining the Effective Width

279
1.5

0.5

case 1 -

1

case 1 -2

b = tc+d b = tc+2d B(Overall Effective width "^'"^^^

1.5 r

>

0.5 b = tc+d b = tc+2d B (Overall Effective width width)

Figure 9. Mcy/ Meu (Maximum Loading) Figure 10. Mcy/ Mey (l (5Case M)

0 10 20 30 40 50 60 70 80 90 100 1 10 120

Displacement (mm)

Figure 11. Load — Displacement Curve

^1

Figure 12. Crack Pattern (Case 2)

280
A Beam modelCj^ *

Beam modelf^)

Pier-footing model

A„

,

_ _

A A

0.0 0.5 1.0 1.5 2.0 2.5 3 0 Shear span ratio

Figure 13. Se / Sc - Shear Span Ratio

Photograph 1 State of the Failure (Case M)

281
Photograph 2 State of the Failure (Case 2)

282

J
DYNAMIC CHARACTERISTICS OF MAGNETO-RHEOLOGICAL DAMPER

by

Hideo FTJJITANI", Katsuaki SUNAKODA^'. Hiroshi SODEYAMA'', Norio IWATA^' and Satsuya SODA^'

ABSTRACT

Two kinds of magneto-rheological fluid damper

(MR damper) have been designed and

manufactured. One has a nominal capacity of 2kN

and the other 20kN. A bypass flow system is adopted for both dampers and each has the same

capacity of electromagnet attached to the bypass

portion. The effective fluid orifice is the

rectangular space and the magnetic field is applied

from the outside.

A test was performed by applying different

magnetic fields to the orifice portion of the

rectangular space. The damping force and the

force-displacement loop were evaluated

The test results yielded the following : 1) Two

type's of dampers functioned by using one unit of

the electromagnet under an appropriate electrical current control. 2) The magnitude of the damping

force depends on the input magnetic field, but it has an upper limit. 3) Without an applied

magnetic field, the MRF damper exhibits viscouslike behavictt", while with a magnetic field it shows

friction-like behavior.

A mechanical model of the damper is estimated by taking account of the force- displacement loop.

It is clarified that MRF dampers provide a echnology that enables effective semi-active

control in real building structures.

Keywords: Damping characteristic

Magneto-rheological fluids MR dampers

1. INTRODUCTION

Vibration control systems are mainly divided into

three types: passive, active, and semi-active. The

semi-active type attracts a great deal of attention

because of it's the following superior properties:

1) It is basically stable in contrast to the active

type. 2) Its control effect is more flexible than the passive type. 3) Its external energy requirements

are orders of magnitude smaller than those of

typical active types.

Until now, various kinds of semi-active control

devices have been developed. One fundamental

device for semi-active control systems is the

capacity adjustable viscous damper. Its viscosity

is adjustable in accordance with the opening rate of the flow control valve inside the damper. A

series of studies with this type of damper has been

carried out and they wctc found to have good

characteristics and efficient control system

propoties However, magneto-rheological

(MR) and electro-rheological (ER) fluid dampers

attracted significant attention. The damping

properties of these types of dampas are changed

by changing the magnetic/electrical field applied

to each fluid Research on semi-active control of building

structures using controllable fluid dampers is being conducted in the U.S. -Japan coopCTative

research and development project of "Smart

Materials and Structural Systems." launched in 1998 by NSF and Building Research Institute,

Japan Ministry of Construction . This paper

ouflines the results of experimental tests and

simulations of magneto-rehological fluid dampers

conducted as one of the research items of this cooperative research on the Japan side. Two kinds of Magneto-rheological fluid

damper (MRF dampCT) have been designed and

manufactured One has a nominal capacity of 2kN

and the other 20kN. A test was performed by

applying different magnetic fields to the orifice

portion of the rectangular space. The damping

force and the force-displacement loop were

evaluated.

1) Head, Performance System Division, Codes and

Evaluation Research Center, Building Research

Institute, Ministry of Construction, Tsukuba-shi,

Ibaraki-ken, 305-0802, Japan

2) Sanwa Tekki Corp., Shinagawa-ku, Tokyo, 140- 8669, Japan

3) Department of Architecture, Waseda University,

Shinjuku-ku, Tokyo, 169-8555, Japan

283
Furthermore, an analytical model is proposed

to simulate the behavior of the MR dampers, and

simulation and experimental results are compared^

2. DESIGN OF MRF DAMPER

The bypass flow type MRF damper has been

developed in this study. Fig. 1 shows its hydraulic

circuit. The bypass flow portion is a passage for MR fluids connecting two pressure chambers.

The bypass flow is at the outside of the cylinder.

In bypass type hydraulic dampers such as this, a cylinder is divided to two airtight pressure

chambers by a piston with rubber 0-rings. The

MR fluid flows fi-om a high-pressure chamber to a low-pressure chamber through the bypass flow

portion. The bypass flow portion has an orifice

and the MR fluid flow is narrowed rapidly at the

orifice. Moreover, intense magnetic fields are

Re se rvo 1 ^3

r

Co 1 I Figure 1 Hydraulic Circuit of Bypass Type MRF Damper

j&<vW:

Figure 2 MRF Damper - 2kN Figure ^ MRF Damper - 20kN

284
applied to the MR nuid at the orifice by

electromagnets. To simplify the analysis of the

MR fluid flow state in the magnetic field, a rectangular ctoss section is selected as the orifice shape. The magnetic fields are applied

perpendicularly to the MR fluid flow at the orifice. The electromagnet is formed by copper wire

wound around a C-shaped iron rod. The air gap

between the ends of the C-shaped iron rod forms

the rectangular cross section of the orifice. Thus,

the length of the air gap of the electromagnet is equivalent to the thickness of the orifice that is penetrated by the magnetic flux. This method of

magnetizing MR fluids has the advantages that the

MR fluid is rarely affected by the heat generated on the electromagnet.

Two different capacities of MRF dampers have

been designed and built. Fig. 2 and Fig. 3 show

photographs of them. Fig. 2 shows the one whose

the nominal maximum damping force is 2kN. and

Fig. 3 shows the one whose the nominal maximum

damping force is 20kN. Table 1 shows their design

specifications. These two dampers have the same

basic mechanism. In particular, their electromagnets are identical. The electromagnet's

design is based on the larger dampers. It is therefore used under the small input electric currents for the 2kN MRF damper. The basic

specifications of the electromagnet are also shown

in Table 1 - 1 . Very low carbon steel is used for the

C-shaped iron rod to obtain a high level of

magnetic saturation and low residual

magnetizaUon. The 2kN MRF damper's orifice is 0.6 X 16mm in cross section and lOmm long. The

201cN damper's orifice is 2 x 20mm in cross

section and 20nim long. When designing the

orifice size, the Reynolds number of the fluid flow

at the orifice was considered to be less than the

critical Reynolds number.

As shown in Fig. 1, these MRF dampers have

two check valves on the passages to the reservoir.

These check valves are open under the nonvibration condition. Therefore, the thermal

expansion due to the temperature rise of the MR

fluid is absorbed by the reservoir. However, the

check valves are shut during vibrations, and the

flow of MR fluid occurs only through the bypass

portion.

Two diffCTent kinds of MR fluids were tested on the 2kN MRF damper, in order to find the most

suitable commCTcial MR fluid for the bypass type

MRF damper. One was LORD MRF-132LD and

the other was LORD MRF-128NB. Based on the

expCTimental results of these 2kN dampers, MRF132LD was selected as the fluid for the 20kN

MRF damper.

Table 1 Design Specifications of MRF Dampers

MRF damper - 2kN MRF damper - 20kN

Max. Force

(nominal)

2kN 20 kN

Stroke ±10mm ±35 mm

Cylinder bore 35 mm 95 mm

Orifice size 0.6 mmX 16 mm 2 mmX20 mm

Orifice length 10 mm 20 mm

MR fluid (I) MRF-132LD

(D MRF-128NB

MRF-132LD

Coil <l> 0.5 mm 3800 turns

Inductance 1.5 henries

Coil resistance 60 ohms

Max. current 0.08 A 0.16 A

285
3. EXPERIMENT

Various tests have been carried out using of the

vibration-testing machine to verify the damping

characteristics of the developed MRF dampers.

Fig.4 shows a schematic diagram of the

experimental setup for the MRF dampers. The

actuator of the vibration-testing machine has a maximum dynamic load of 2000kN and a maximum velocity of 40cm/s. Various sinusoidal

displacements are applied to the dampers and the

generated damping forces are measured by a loadcell on the opposite side of the actuator. The

input electric current to the electromagnet is

selected as the one of the test parameters and is maintained to a constant value during the dynamic

loading test. Fig.5-1, Fig.5-2 and Fig.5-3 show the

measured force-displacement loops for the 2kN

MRF dampers under maximum input velocities of

5cm/s, 9cm/s and 20cm/s sinusoidal waveform

movement, respectively. The right hand side

loops of every figure show the test results for the

MRF damper using the MR fluid, MRF-132LD,

and the left hand side loops show the results for the MRF- 128 damper. The dynamic loading tests were performed under input electric currents of

OA, 0.0 16 A, 0.032A, 0.048 A, 0.064A and 0.08 A.

Da t a Recorder

A

S I gna

1

P rocesso r

286
-25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm]

3000

2500 2000

1500

1000 500

0 -500

-1000

-1500

-2000

-2500

-3000

2 Hz

0 08 A

0 084 A

0 048 A

0 032 A

0 016 A

0 A

( MRF-128NB )

-25 -20 -15 -10 -5 0 5 10 15 20 25

Displacement [mm]

Figure 5-1 Force-displacement Loops ai 5cm/s

-25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm]

o

-25 -20

Displacement [mm]

Figure 5-2 Force-displacement Loops at 9cm/s

-25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm]

o

-25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm]

Figure 5-3 Force-displacement Loops al 20cm/s

287
3000

2500

2000

1500

1000 500

0 -500

-1000

-1500

-2000

-2500

-3000

+ 0.08A 0.064A A 0.048A

K 0.032A • 0.01 6A O 0 A

1 1

' '

1 1 1

(a)

MRF-132LX)

-25 -20 -15 -10 10 15 20 25

3000

2500

2000

1500 1000

500

0 -500

-1000

-1500

-2000

-2500

-3000

+ 0.08A 0.064A A 0 048A

X 0032A • 0.0 16A 0 0 A lilt 1 1 1 1

(b) . MRF-128NB

-25 -20 -15 -10 -5 0 5 10 15 20 25

Velocity [cm/s] Velocity [cm/s]

Figure 6 Force-velocity Relationship for 2 kN MRF Damper

z

20

15

10

5

0

-5

-10

-15

-20

0.5Hz ( a )

0.16

0.08

0 A

-25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm]

z

20

15

10

5

0

-5

-10

-15

-20

0.5Hz

0.16

0.08 0 A

(b)

-25 -20 -15 -10 -5 0 5 10 15 20 25

Displacement [mm]

-25 -20 -15 -10 -5 0 5 10 15 20 25

Displacement [mm]

-25 -20 -15 -10 -5 0 5 10 15 20 25

Displacement [mm]

Figure 7 Force-displacement Loops for 20kN MRF Damper

288
It was verified that the maximum (lamping

force was controllable by adjusting the magnetic

field. Furthermore, an upper limit of the

controllable damping force was observed. The

magnetic field of the electromagnet can be

increased linearly to approximately 0. 16A.

Therefore, it is understood that this phenomenon

is due to the magnetic saturation of the MR fluid. The force-displacement loop under no magnetic

field exhibits the behavior of a viscous damper,

especially in the high velocity range of around

20cm/s. The friction force of the sealing or the

damping force due to the residual magnetization is dominant for the total damping force in the low

velocity range. Fig.6(a) and Fig.6(b) show the force-velocity

relationsWps. Although the absolute values of the

maximum damping forces of the two dampers are

similar, the increase rate of the damping force of

the MRF damper using the MRF-128NB is higher

than that of the damper using the MRF-132LD.

The dynamic range of the damper using the MRF128NB is smaller in the high velocity range. These results mean that the MRF-132LD is more

suitable for the developed bypass type MRF

dampers. The MRF-132LD was used for the

20kN MRF damper

Fig. 7 shows the force-displacement loops for the 20kN MRF damper under 0.5Hz sinusoidal

waveform. Fig. 8 shows the relationship between

the maximum input velocity and the measured

maximum damping force. It was verified that the

electrical currents applied to the electromagnet

controlled the damping force of the 20kN MRF

damper. The characterisUcs of the damping forces

at the low velocity level, less than 5cm/s, were

also invesfigated through these tests. It was

confirmed from the Fig. 7(a) and Fig. 8 that the MRF damper functioned under a comparatively

low velocity and over a small displacement range.

289
4. ANALYTICAL MODEL

FORMULATION

This section proposes the analytical model for simulating the behavior of the MRF dampers, and

compares the simulation and experimental results. Until now, some types of mechanical models for MR(ER) fluid dampers have been proposed. The

simplest one is the Bingham viscoplastic model,

in which a couple comprising a dashpot and a friction slider are connected in parallel. Gavin

used the mechanical model proposed by Gamota

to simulate the behavior of ER fluid dampers, in which the Zener element shows frequency

dependent behavior over a wide range of the

frequencies [Gavin et al. (1998)] ^'

. The model

proposed by Spencer consists of two springs, two

dashpots and the Bou-Wen model, which can

exactly simulate the behavior of both the

displacement-force and velocity-force

relationships of MRF dampas [Spencer et al. (1997)] ^'

.

In this paper, the involution model is used to simulate the behavior of MRF dampers. In this model, the velocity-force relationship is expressed

by:

F=CV (1)

where F is the damping force, C is a constant independent of the frequency; V is the velocity of

the piston and a is an exponent such that 0.0 < a

<1.0. This expression has often been used to simulate viscous fluid dampers, and is available to simulate the behavior of MRF(ERF) dampers

because the damping force remains within the

specified bound under the condition that a is close

to zero. In addition, since there are only two

parameters in this expression(C and a), they are

easier to estimate than the elements in the mechanical models mentioned above.

Fig.9 and Fig. 10 shows the simulation results of the 21cN damper obtained from Expression (1). Both parameters(C and a) are assumed to be

independent of the amplitude and the frequency.

Values of C and a, shown in Table2, were

determined to minimize the square of the errors between the experimental force and the analytical one. Good agreement with experimental results is confirmed for each applied current. The experimental results of the 20kN damper

were simulated by following the same procedure

as for the 2kN damper. In this case, though, ideal sine waves were used as input displacement

because some noises were included in observed

displacement. The results are shown in Fig. 11. Values of C and a are identified as shown in Tables. As well as the case of the 2KN damper, it

is found that the analytical model can exactly

predict the behavior of the experimental results. Through a series of simulations, it is confirmed

that the behavior of MRF dampers is exactly

predicted by the velocity-force relationship

expressed in (1) over a wide range of applied

current, amplitude, and frequency. However, it should be noted that this analytical model can be

applied only to simulate the dynamic behavior of

MRF dampers, because the intrinsic property of

the term "CV"" can not simulate the static behavior.

290
-25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm]

-25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm]

(a) MRF-132LD

Figure 9 Simulation Results (2lcN Damper, Frequency=2.0Hz)

(b) MRF-128NB

-25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm]

-25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm] (a)MRF-132LD

Figure 10 Simulation Results (2kN Damper, Frequency=3.0Hz)

(b) MRF-128NB

Table 2 Values of Analytical Model {F=C V) for 2kN Dampers

Current[A]

MRF-128NB MRF-132LD

C\N(mn}/sn a C[N('m/7z/.y)"l a

0.000 91

A

0.47 85.6 0.40

0.016 349 0.31 292 0.26

0.032 608 0.23 576 0.20

0.048 788 0.20 704 0.19

0.064 940 0.18 752 0.19

0.080 920 0.19 812 0.18

291
0 16 A

0.08 A

0 A

_J L.

25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm]

(a) Amplitude=2.5[mm]

-25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm]

(b) Amplitude=5[mm]

0 16 A

0.08 A

25 -20 -15 -10 -5 0 5 10 15 20 25 Displacement [mm]

(c) Amplitucle=12.5[mm]

20

15

10

5

V 0

o

o

11. -5

-10

-15

-20

J I 1_

\ 0.16 A

\d.OS A

-25 -20 -15 -10 -5 0 5 10 15 20 25

Displacement [mm]

(d) Amplitude=20[mm]

Figure 1 1 Simulation Results (20kN Damper, Frequency=0.5[Hz])

Table 3 Values of Analytical Model (F=C V^) for 20kN Damper

Current[A] C[KN(mni/sy a

0.00 0.065 0.91

0.08 2.59 0.28

0.16 3.84 0.28

292
5. CONCLUSIONS

Various tests have been carried out using a vibration-testing machine to verify the damping

characteristics of developed MRF dampers. The

following test results were obtained: (1) Two

types of dampers functioned by using one unit of

the electromagnet under an appropriate electrical current control. (2) The magnitude of the

damping force depends on the input magnetic

field, but it has an upper limit. (3) In the absence

of an applied magnetic field, an MRF damper

exhibits viscous-like behavior, while it shows

friction-like behavior in a magnetic field. Through a series of simulations, it is confirmed that the behavior of the MRF dampers

is fairly predicted by the velocity-force

relationship expressed in equation (1) over a wide

range of applied current, amplitude, and frequency.

It is clarified that the MRF dampers provide a technology that enables effective semi-active

control in real building structures.

ACKNOWLEDGMENTS

The authors would like to express their appreciation to the members of ER/MR working

group in the Effector section of the Japan side of

the project, "Smart Materials and Structural

Systems" for their fruitful discussions.

REFERENCES

1. Niwa, N., Kobori, T., Takahashi, M.,

Matsunaga, Y., Kurata, N. and Mizuno, T. (1998). Application of Semi-active Damper

System to an Actual Building. Froc., 2nd

World Conference on Structural Control,

pp.8 15-824.

2. Kurita, N., Kobori, T., Takahashi, M. and

Niwa, N. (1998). Semi-Active Damper

System in Large Earthquakes. Proc., 2nd

World Conference on Structural Control,

pp. 359-366.

3. Kawashima, K., Otsuka H., Unjoh S. and

Mukai, H. (1995). Development of Hybrid

Seismic Response Control oi Bridges. Prix:.. 27'^ Joint Meeting of UJNR, pp.295-316.

4. Gavin., H.P. and Hanson, D. (1998). Seismic

Protection using ER Damping Walls, Proc.,

2nd World Conference on Structural Control,

pp. 1183-1 190.

5. Hidaka, S.. Ahn, Y.K. and Morishita, S. (1998). Structural Control by DampingVariable Dynamic Damper using ER Fluid. Proc., 2nd World Conference on Structural

Control, pp.45 1-460.

6. Fukukita, A., Tamura, K., Hayashi, S. and

Shiba, K. (1995). Semi-active control of

rotauonal variable damper using elctro- rehological fluid. Journal of Structural

Engineering, Vol.41B, pp.23-32. (in Japnese)

7. Spencer Jr., B.F., Dyke, S.J., Sain, M.K., and

Carlson, J.D. (1997). Phenomenological

Model for Magnetorheological Dampers. J. of

Engineering Mechanics, ASCE, Vol. 123,

No.3, pp.230-238

8. Johnson, E., Ramallo, J.C., Spencer Jr., B.F. and Sain, M.K. (1998). Intelligent Base

Isolafion Systems. Proc., 2nd World

Conf^ence on Structural Control, pp.367-376

9. Carlson, J.D. and Spencer Jr., B.F. (1996).

Magnelo-Rheological Fluid Dampers for Semi-Active Seismic Control. Proc., 3rd Int. Conf. on Motion and Vib. Control, Vol.3,

pp.35-40.

10. Spencer Jr., B.F., Yang, G., Carlson, J.D. and

Sain, M.K. (1998). "Smart" dampers for Seismic Protection of Structures: A Full-Scale

Study. Proc., 2nd World Conference on

Structural Control, pp.4 1 7-426.

11. Jolly, M.R., Bender, J.W. and Carlson, J.D. (1998). Properties and Applications of

Commercial Magnetorheological Fluids. Proc. SPIE's 5th Annual International Symposium

on Smart Structures and Materials.

12. Otani, S., Hiraishi, H., Midorikawa, M.,

Teshigawara, M., Saito, T and Fujitani, H.

(2(X)0). Development of Smart Systems for Building Structures. Proc., SPIE's 7th Annual

International Symposium on Smart Structures and Materials.

293
i
Theme 4

Performance-Based

Design Methods

295
I
COMPARISON OF PERFORMANCE REQUIRED BY BRIDGE DESIGN CODES

IN VARIOUS COUNTRIES

by

Michio OKAHARA Jiro FUKUI '\ Masahiro NISHITANI and Kenji MATSUI "

ABSTRACT

The authors carried out a comparative analysis

of the regulations for bridge design codes in various countries, as for required performance

and seismic performance etc. In Japan, the

revision of the Design Specifications of

Highway Bridges to performance based design

is now under study. This comparison analysis

will make a significant contribution to the

revision of the Specifications. The end of this report concludes with a summary of principal contents and an outline of future issues.

KEY WORDS: Bridge Design

Performance Based Design

Required Performance

Seismic Performance

1. INTRODUCTION

Around the world, the adoption of

performance based codes is being debated by

many organizations in architectural field and

has been partially completed. In Japan, new

Building Standard Law, that is performance

based standard, will be enforced the beginning

in 2000 and a trial of performance based

ordering of paving work is being carried out. Performance based design clarifies the

performance that structures require, and a number of suitable verification methods to be

used to verify their performance have been

proposed.

In Japan, the revision of the Design

Specifications of Highway Bridges to performance based code is now under study.

A comparative analysis of the performance

required by bridges under codes in various

countries will make a significant contribution

to this study.

To study the requirements for performance

based design in bridge design codes, the

authors performed a comparative analysis of

the regulations for general bridge design,

required performance and seismic

performance in present bridge design codes in various countries. Excluding the articles specifying seismic design in the codes of

various countries, there are no complete codes

that can be defined as performance based

codes. Consequently, the authors focused and

arranged at terminology such as "philosophy"

and "requirements" in various countries'

codes.

The study dealt with the following codes.

1) Director, Structure and Bridge

Department, Public Works Research

Institute, Ministry of Construction,

Tsukuba-shi, Ibaraki-ken, 305-0804,

JAPAN

2) Head, Foundation Engineering Division,

ditto

3) Senior Research Engineer, ditto

4) Chief Engineer of R&D Headquarters,

CTI Engineering Co., Ltd., Chuo-ku,

Fukuoka-shi, 810-0041

297
Regulations and required performance for general bridge design compared were those

stipulated in the LRFD version of the

AASHTO, Eurocode, Caltrans and the revised

edition of the Design Specifications of

Highway Bridges (Japan, now under study).

And the seismic design in the LRFD version

of the AASHTO, Eurocode, Transit Bridge

Manual (New Zealand) and Design

Specifications of Highway Bridges were

compared.

(1) United States of America;

- AASHTO LRFD Bridge Design

Specifications (1998, 2nd Edition)"

[AASHTO]

(2) European Union;

- Draft prEN1990: Basis of Design (1999)'*

- Preliminary Draft EN1997-1: Geotechnical

Design, Part 1 General Rules (1999)"

[Draft EN]

- ENV1998: Design provisions for earthquake

resistance of structures (1994)" [ENV1998]

(3) State of California, U.S.;

- Memo to Designers 20-1 /Seismic Design

Methodology (1999)"

- Caltrans Bridge Design Specifications

(1990)''

- Seismic Design Criteria, version 1.1(1999)^'

[Caltrans]

(4) New Zealand;

- TNZ Bridge Manual for Design and

Evaluation (1995)'* [TNZ]

(5) Japan;

- Revised edition of the Design Specifications

of Highway Bridges (200X)'* '°' [Draft JRA]

2. OVERVIEW OF PERFORMANCE

BASED DESIGN

The term of performance based design is interpreted in various ways. A broad

categorization of codes that are called

performance based design places the

following two classes.

In one case, following the arrival of the age

of mega-competition as a consequence of the

appearance of a single world market after the

collapse of the Soviet Union that ended the

cold war, the international community needed

common rules. And they adopted

"performance" as a key word accepted by all nations. The architectural field has past

provided many performance elements

including seismic, safety, fire resistance and

human-friendly performance. And constraints

applied by regulations and laws by

governments on these various performance

elements are obstacles to international trade. In the Nordic countries, a movement to

establish "performance" as a key word that would destroy these obstacles began in the

early 1970s. This movement spread to England in the 1980s, to Australia and New

Zealand in the 1990s, and in recent years has

been reflected in the revision of the Building

Standard Law of Japan. So based on the

hierarchical character of regulations (or

required performance), the degree to which

they are endowed with legally constraining is debated. Figure 1 presents a hierarchy of

regulations based on the Nordic 5 level system'". The only legal objection is generally

whether or not a building's performance is guaranteed and designers are given greater

fi-eedom to decide on the method to verify

their performance. In this case, performance

based design can be interpreted as synonymous with ft"ee design method.

In the second case, the need for developed

seismic technology has been pointed out and

great progress has been achieved based on the

lessons learned from a series of earthquake

disasters beginning with the Loma Prieta

(1989), Northridge (1994) and Hyogo-ken

298
Nanbu Earthquake (1995). Because, partly as a result of its technological difficulty, the

limits to the capacity of structures against a

large earthquake were not adequately

explained to owners (or users) who were not

professionals, they fell into the illusion that most buildings could withstand any

conditions. Awareness of this problem

triggered a movement to clarify performance

as a design index that can gain the mutual

approval of owners and designers. Vision

2000'^' that the Structural Engineers

Association of California (SEAOC) rushed to codify performance that structures require in the form of a performance matrix. Figure 2

shows an example of a performance matrix.

The horizontal axis is called performance

level, that is corresponding to performance

(equal to limit states), and the vertical axis is called hazard level.

These movements appeared first in the

architectural field, and it can be stated that at this time, codes incorporating these

approaches have not been established in bridge design field anywhere in the world.

Since the Northridge and Hyogo-ken Nanbu

Earthquake, performance requirements related

to seismic performance has been carried out based on the results of technology

development, but nothing has been done

outside of the seismic design field. Therefore,

a comparison of required performance for

bridge design had been carried out by

focusing and arranging design philosophies or requirements in various codes.

3. COMPARISON OF PERFORMANCE

REQUIRED AND OTHERS BY BRIDGES

Table 1 presents required performance,

foundation design, etc. that are stipulated in codes for four regions, countries, or states.

The followings are special mentions.

- Because the characteristics of the codes

differ, required performance of bridges is stipulated from their respective perspectives.

- Because the AASHTO and Draft JRA deal

directly with bridges, the required

performance is specific.

- Eurocode is not classified by structural form; rather by each material used in civil engineering and architectural fields.

Consequently, the Draft EN does not directly

specify required performance for bridges.

- Caltrans stipulates the basis of seismic

design on the premise that it will be applied

to "normal or standard bridges." The

regulations excluding seismic provisions of

the Caltrans are fundamentally correspondent

to the AASHTO.

- Draft JRA is based on a proposed revision

that is now under study. Consequently, as in the case of the Draft EN, finally its configurations and contents may change.

(1) Required Performance

As for the Draft EN, the strict required

performance can not be described, because a

structure has not been specified on the

character of the code. Therefore, in the Draft

EN, the general limit states are stipulated,

substituting with the concrete required

performance. In the Draft EN, ultimate limit

state and service limit state has been set. The

former limit state also contains not only

structural safety but also human safety, and

appearance is also included in the latter limit

state.

The required performance of the Caltrans for

large earthquakes for bridges is simple and

concrete, and only collapse limit state of a

structure, this is a kind of ultimate limit state,

is included. In the AASHTO, economical

efficiency, beauty and landscape are

299
emphasized in addition to safety and

serviceability of structures. To ensure the

safety of bridges, ductility, redundancy and

importance are considered in the code. And,

the performance of serviceability is used for

the wide meaning, including durability,

inspectability, maintenanceability, amenity and

deformability, etc.

In the Draft JRA, it would be specified almost

equal required performance of the AASHTO.

As an examination of economical efficiency,

it would be epoch-making to recommend the

evaluation of life cycle cost of a bridge. Still,

in the design of substructures and seismic

design of the Draft JRA, performance I~III

and seismic performance I~ III would

concretely be set as required performance for the safety of bridges.

The design of bridges which harmonize with

beauty, landscape and environment would be

emphasized, as shown in Table 1.

(2) Basis of Design

Ductility, redundancy and importance have

been taken up as the basis of bridge design.

In all codes, ductility in seismic design has

been evaluated. In the AASHTO and Draft

EN, redundancy as a bridge structure has been

noticed, and in the Caltrans and Draft JRA,

redundancy as a road network after

earthquake has been noticed. Importance has

been applied to seismic design for all codes.

(3) Design Service Life

A life is distinguished by service life and

design life. Service life is correspondent to what is called design service life. Design life

is correspondent to reference period of

IS02394(1998)'", and design life is a period

as a base for calculating fluctuating loads.

In the Draft EN and Draft JRA, the design

service life of bridges is shown as 1 00 years. On the other hand, in the AASHTO, service

life has been not specified, generally it has

been understood with 100 years. Still, design

life has been specified as 75 years.

(4) Life Cycle Cost

In the AASHTO, life cycle cost has been not

described. In the U.S., the Life Cycle Design

and Performance (1997)"" that has been

prepared by the U.S. Army Corps of

Engineers is the regulation that stipulates the

basic framework for performance based

design applied to public works projects. It stipulates that engineering decisions should

not do simply minimize initial costs or maximize reliability that ignores costs, and

demands that designers perform evaluations

based on life cycle cost of an entire project.

The design of highway bridges in the U.K. is based on BS5400 (Code of Practice for Bridges, BSI) and conforms with many

Government Departmental Standards in the

Design Manual for Roads and Bridges

(DMRB). The code BD36 in the DMRB

provides the method of comparing life cycle

cost (the design service, life in the U.K. is 120

years premised on appropriate maintenance).

The Draft JRA is the first code to clearly

stipulate the concept of evaluating the

optimum proposal based on total cost during

service life.

(5) Limit States Limit States are engineering indices for checking whether structures satisfy their

required performance. In the AASHTO,

service limit state, fatigue limit state, strength

limit state and extreme event limit state are

set. Each limit state means the following;

- Service limit state: under usual service

conditions, stress, deformation and crack are

300
limited.

- Fatigue limit state: stress width by repetition

traffic loads is limited.

- Strength limit state: for load combinations

which encounter during design life, the

strength and stability of the local and whole

of a structure are compensated for. Though

damages are received, the whole structural system is maintained.

- Extreme event limit state: for a large

earthquake, flood and the collision of a ship,

the residual capacity of a bridge is guaranteed.

The basis of seismic design of the Caltrans is based on the deformation. The purpose of

seismic design is to check the deformation

occurring by a large earthquake, and in the

Caltrans, only collapse limit state is specified.

Collapse limit state of a structure is defined as a state not resisting to its dead weight, by

generating large deformation.

In the Draft JRA, limit states are not

stipulated, because of using an allowable

stress design (WSD) method.

(6) Basis of Verification Method

The verification equation of a limit state in the AASHTO is basically provided as the

following.

E 7? i 7 .Qi ^ 0 R«= Rf

Where:

77 ,= 7] D 7) R 77 1^ 0.95,Or 71 ,= !/( 7? D 7] R 7] i)^ 1.0 7 .: load factor

Q.: load effects

<t> : resistant factor (factor that is multiplied

by nominal resistant force)

Rn-. nominal resistant force

Rr: resistant force multiplied by resistant factor (design resistant force)

Here, the factor t] d for ductility and the factor

77 R for redundancy are used to verify strength

limit state, and tKe factor 77 i for importance is used to verify extreme event limit state. As

the AASHTO, in future bridge design, the

effects of ductility and redundancy must be

represented by a factor in some way in codes.

The design method of the Caltrans converted

from conventional force based design method

to displacement based design method. In seismic design, it seems to introduce by and

by this design method.

(7) Load

As shown in Table 1, the classification of

loads is different by each code at present. The

classification of the Draft EN is fimdamentally similar to the regulations of

IS02394. Hereafter, it is required to match

the ISO code.

In the Draft JRA, the effect of salt attack, effect of corrosion, fatigue by the repetition of

live loads would be newly considered as other

actions and effects.

(8) Foundation Design

In the AASHTO, three limit states except for fatigue limit state have been checked. In the

calculation of resistance force (design vertical bearing capacity), resistant factor 0 1 vis considered. Here, 0 is the factor which

depends on the estimation method of bearing

capacity, and X v is the factor which depends on construction management and estimating

system of bearing capacity. In the Eurocode,

such resistant factor is not given. The desired

value of the safety index jS in calculating

this resistant factor is 2.0 ~ 3.0, according to Appendix A'" of Barker et al (1991).

The foundation design of the Draft EN has

been specified at EN 1 997-1 (Geotechnical

design). The MFA (Material Factor Approach)

and RFA (Resistance Factor Approach) have

been proposed as partial design formats of

301
ground resistant in Draft EN. The design

resistant force Rd based on each approach is represented by the following equations.

Approach of MFA: Rd = R(Xr/ y m, a^or) Approach of RFA: Rd = R(Xr, a^or)/ 7 r / ^ Where:

Xr: characteristic value of ground materials

7 m: material factor

7 r: resistant factor

anor: nominal value of geometric quantity ^ : correlation ratio In the MFA approach, the partial factor 7 m, as shown in Table 1, is directly considered in ground parameters. The RFA approach is used

for the design of piles, and it is similar to the

LRFD approach. The correlation ratio ^ is a

factor considering the frequency of in-situ loading tests.

In the Caltrans, both of SLD (Service Load

Design) and LFD (Load Factor Design)

method are ever used as design methods. The

resistant factor 0 has been applied to the

latter method. Here, the 0 is divided in every Groups (combination of loads), and in the Group VII (present SDC) including an earthquake action, factor of 1.0 is applied, and

in other Groups, factor of 0.75 is applied.

There is no relation between ground

parameters and estimation methods of bearing

capacity.

In the present JRA, because the reliability of

bearing capacity of a pile by in-situ loading

tests is higher than that of empirical

estimation method, the correction factor 7 is applied to estimate bearing capacity to reflect the effect it. And, the introduction of the

factor which represents quality and quantity of

ground parameters has been also examining as

future problem.

4. COMPARISON OF SEISMIC

PERFORMANCE

Table 2 shows the comparison of seismic

regulations in various codes. Many studies

such as that by Backle (1996)"* or NCEER

(1997)'*', have compared and considered

regulations in seismic design codes in various

countries. This report adds considerations

from the perspective of international

harmonization and locality of codes. The New

Zealand's TNZ Bridge Manual (1994) instead

of the Caltrans is shown in Table 2.

(1) Basis of Seismic Design

The AASHTO clarifies seismic performance

for two levels: medium earthquakes and large

earthquakes. The ENV1998 and TNZ mention

to traffic ftinctions after an earthquake. The

present JRA on the other hand treats differences in seismic performance required

according to the importance of a bridge as

basis of seismic design.

(2) Seismic Performance

Because in the AASHTO, the effects of

earthquakes consider only extreme event limit

state as stated in Table 1, the seismic

performance required has been interpreted as only this limit state. As shown in 1) of Table

2, the AASHTO has adopted a two-level

design seismic force, but it is not clear to the

authors whether or not, in practice, this extreme event limit state is verified for both

medium and large earthquakes. The limit

states in the Eurocode are basically service

limit state and ultimate limit state. In conformity with this concept, the ENV1998

applies these two limit states to

post-earthquake service and post-earthquake

safety respectively, with the former verified

based on displacement and the latter verified

based on strength of a structure. In the TNZ,

it stipulates, not clarified in the Bridge

Manual, that an earthquake equivalent to an

302
ultimate limit state with return period of 450

years be considered'". Based on the

earthquake motion hypothesized by this level of earthquake, the seismic performance is specified for small and large cases. The

present JRA combines three limit states, post-earthquake service, post-earthquake

restoration and post-earthquake safety with

bridge importance A and B. Judging from

bridge importance, two seismic performances,

post-earthquake service and post-earthquake

restoration (for a case of class B bridges), and

post-earthquake service and post-earthquake

safety (for a case of class A bridges) are combined with level 1 and level 2 earthquake

motion respectively.

(3) Importance

Although definitions of bridge importance

differ, they are categorized in two or three

classes in all codes. Categorizations based on

importance are, in the case of the AAHSTO,

reflected in the factor for importance r? i in the calculation equation for the design load Q.

Importance in the ENV1998 and TNZ is related to adjusting the return period of design

earthquake motion for a normal bridge, but

for practical design procedures, the design

earthquake force of a normal bridge is increased or lowered. In the case of the

present JRA on the other hand, it is reflected

in differences in the required seismic

performance indicated in h) and in differences

in the safety factor used to decide an

allowable ductility factor. Considered from

this perspective, in contrast to the AASHTO,

ENV1998 and TNZ, the specifications of the

JRA are detailed.

(4) Zone Classification

The AASHTO has set the value A for a

construction site based on the hazard map of

return period of 475 years (10% probability of

exceedance during service life of 50 years).

The seismic zone has been set as class 1 to class 4 according to the value of A. This zone

factor has been used to establish differences

in required analysis methods, detailed of

bridge piers, and bridge abutments and

foundations design method. The ENV1998

has not set specific zone factors, but the

design ground acceleration ag reflects the

degree of earthquake risk in each nation. The

TNZ has provided the contour map from 0.4

to 0.8 as its zone factor Z. In New Zealand,

instead of the present deterministic hazard

map, the distribution map of maximum

ground acceleration based on a probabilistic

risk analysis would be used as the zone factor

(Figure 3).

(5) Design Earthquake Motion

The AASHTO has set two design earthquake

levels. For a medium earthquake, it is based on the value of A for return period of 475

years (10% probability of exceedance during

design service life of 50 years). For a large

earthquake, the maximum credible earthquake

(MCE) is considered to be an earthquake with

return period longer than the design

earthquake: an earthquake with return period

of 2,500 years for example. The ENV1998

defines a large earthquake as an earthquake

with return period of 475 years, but regarding

a medium earthquake, it simply stipulates:

"earthquake activity with an appropriate return period shall be evaluated." In the provisions

of the TNZ, one level earthquake motion of

return period of 450 years is considered. The

present JRA defines a medium and large

design earthquake motion as earthquake

motion with a high probability of occurring

during the service life of a bridge and strong one with a low probability of occurring

respectively, but does not stipulate specific return periods.

(6) Elastic Response Spectrum

303
The part m) of Table 2 presents the basic

equation for the elastic response spectrum in each code. Here, only the TNZ uses

non-linear spectrum (yield seismic intensity).

New railway seismic codes issued in 1999, in Japan, use non-linear response spectrum^"'.

(7) Load Reduction Factor The AAHSTO and ENV1998 establish

reduction factors R or q based on the

importance of a bridge and the type of a

substructure as ways to reduce the elastic seismic force accounting for the non-linear

response of ductile members. The TNZ on the

other hand, has not set a load reduction factor

because it has used non-linear response

spectrum for allowable ductility factor ii (=

1 to 6). The present JRA differs from the

other codes in that it applies the law of

energy conservation to evaluate the load

reduction as the reduction of design lateral force.

(8) Over strength Factor in Capacity Design

A designer must guarantee appropriate

differences in the safety (hierarchy of

strength) of non-linear members and ductile

members in order to count on non-linear

behavior of ductile members that the designer

intends. In the design of a foundation, the

design lateral force that a factor is multiplied

by the ultimate lateral capacity of the pier is used. The factor is an over strength factor. The over strength factor is quantitatively

related to the value of q in the ENV1998. The

TNZ only refers to the capacity design

principles in the NZS4203:1992; it does not

specify a specific factor value. Because the

present JRA provisions regarding the

evaluation of the design lateral force used for

the design of a foundation specifies that it must be 1.1 times the ultimate lateral capacity

of the pier, it is generally understood that capacity design is introduced under present

conditions.

6. SUMMARY AND FUTURE ISSUES

The authors have been studying the

performance regulations to compare

specifications for general bridge design and

specifications for seismic design in codes in various countries. This report concludes with

a summary of principal items and an outline

of future issues.

(1) The design methods adopted by the codes

surveyed include LRFD, LSD and WSD

method. They should be unified in order to

clarify the safety of a structure. The ISO code

concerning design methods for structures is IS02394, and the goal is the LSD or LRFD

method of partial factors format that is stipulated in IS02394.

(2) Under present conditions, stipulated contents that define the required performance

for bridge design varies between the codes.

The stipulated contents and required

performance are hierarchical, and it is necessary to accompany debates concerning

the extent to which they provide legal

constraining by unifying them based on the

triangle structure shown in Figure 1.

(3) As in the case of the AASHTO, it is important that in codes the effects of ductility

and redundancy are represented by factors in some way in future bridge design.

(4) The code that refers to life cycle cost is only the Draft JRA. This concept is indispensable for codes that advocates

performance based design. Design service life

is clearly defined by the Draft EN. It is, therefore, necessary to clarify the grounds for

setting the service life 100 years for bridges.

304
(5) The load categorization in the Draft EN

differs from that in the other codes. In the

Draft EN that is based on IS02394 (1998)

and is internationally unified. Future load

categorization and combinations applied in bridge design should comply with the ISO

code.

(6) As for founuauon design, only the MFA

approach of the Draft EN has considered

partial factors for ground parameters. The

relation between the estimation method of

bearing capacity and resistant factor is detailed in the AASHTO. In the Draft EN

(RFA approach) and present JRA, the bearing

capacity with in-situ loading tests has been

estimated, accounting for a resistant factor.

(7) The basis of seismic design is, in many

cases, focussed on serviceability, safety and

inspectability after earthquakes. As pointed

out by Buckle (1996)"*, the future issues are

to quantitatively represent this using

engineering indices.

(8) Regarding importance categorization, the

JRA reflects importance in differences in seismic performance, but in the other codes, it

is linked to differences in the value of design

seismic forces.

(9) Except for the TNZ, the other codes set two levels of earthquake motion. The JRA

actually provides for the performance of two

design levels while the other codes stipulate

verifications against a large earthquake and

practically eliminate verifications against a middle earthquake. A two-level design

method is more rational than one-level. It is necessary to study the matter to determine

whether it is necessary for practical designs

accounting for the importance of bridges.

(10) Excluding the JRA, the others evaluate

earthquake motion probabilistically in the sense that they all account for return period.

ISO/CD3010 (1999)"', going one step fiirther, the uncertainty of earthquake motion based on

probabilistic theory has been accounted as load factors. Because internationally, research

has progressed from conventional MCE based

deterministic risk analysis to probability

analysis, the movement to introduce

probability theory to the evaluation of design

earthquake motion has gained momentum.

(11) At this time, only the TNZ has accounted

for non-linearity in response spectrum. This

method has been adopted in railway bridge

codes in Japan. After the earthquakes in Turkey and Taiwan, the evaluation of active

faults has become an important research

theme. The Caltrans has stipulated that the

spectrum be increased by a maximum of 20

% depending on the distance from a fault.

(12) The AASHTO and Draft EN have used

load reduction factors (R value or q value) to reduce an elastic response value. A particular

characteristics of the latter code is that an over strength factor used for capacity design

has been represented as a function of the q

value. The introduction of LSD method to seismic design is counted on to provide for the evaluation of load factors and evaluation

method of rational over strength factors.

In conclusion, the authors wish to express

their opinion regarding precautions that should

be considered during revision of the present

JRA. As stated at the beginning of this report,

performance based regulations in future codes

is essentially international trends. It is also

necessary to achieve harmonization with

international rules, and IS02394 and ISO3010

codes must be considered as closely as

possible. In this sense, an LSD method based

305
on partial factors format (or an LRFD

method) including seismic design

specifications should be introduced to the

future JRA. On the other hand, differences in the safety level of structures and natural

environments, and regional and national

differences in the stock of design and

construction technology, and the level of

engineers are the matters that must be

recognized as localities, and which differ from

items that should be unified. The revision of

the present JRA must be carried out to achieve performance based design and to consider international harmony and locality.

REFERENCES

1) AASHTO: AASHTO LRFD Bridge Design

Specifications (2nd Edition), 1998.

2) CEN: Draft prEN1990: Basis of Design,

1999.12.

3) CEN: Preliminary Draft EN1997-1:

Geotechnical Design, Part 1 General Rules,

1999.4.

4) CEN: ENV1998: Design provisions for

earthquake resistance of structures, 1994

5) Caltrans: Memo to Designers 20-1/ Seismic

Design Methodology, 1999.1.

6) Caltrans: Caltrans Bridge Design

Specifications, 1995-.

7) Caltrans: Seismic Design Criteria

(Version 1.1), 1999.7.

8) Transit New Zealand: TNZ Bridge Manual

for Design and Evaluation, 1995.

9) Japan Road Association: Draft of

Specifications for Highway Bridges, Part 1: Common Rules, 200X.

10) Japan Road Association: Specifications

for Highway Bridges, Part V: Seismic Design,

1996.

11) NKB: Structure for Building Regulations,

NKB report No.34, 1978, 11. 12) PBSE Ad Hoc Committee:

Force-Displacement Approach for

Performance-based Seismic Engineering - Preliminary Guidelines -, SEAOC Blue Book

Appendix I-PBSE Guidelines: Part 2, 1999.

13) ISO: IS02394: 1998 General principles on reliability for structures, 1998.

14) U. S. Army Corps of Engineers,

Department of the Army: Engineering and

Design LIFE CYCLE DESIGN AND

PERFORMANCE, CECW- ED Engineer

Regulation 1110-2-8159, 1997.

15) AASHTO : Standard Specifications for Highway Bridges, 16* Edition, 1996.

16) NCHRP: Load Factor Design Criteria for Highway Structure Foundations- final report,

NCHRP 24-4, 1991.

17) I. G. Buckle: Overview of Seismic Design

Methods for Bridges in Different Countries

and Future Directions, Proceedings of

IIWCEE, 1996.

18) C. Rojahn et al.: Seismic Design Criteria

for Bridges and Other Highway Structures,

Technical Report NCEER -97-0002, 1997.

19) R. Park: New Zealand Perspectives on

Seismic Design of Bridges, Proceedings of

IIWCEE, 1996.

20) M. W. Stirling et al: Development of a New Seismic Hazard Model for New Zealand,

bulletin of NZNSEE, 1999.

21) Railway Transportation Research Institute: Seismic Design for Railway Structures, 1999.

22) ISO: ISO/CD3010: 1999 Bases for design

of structures- Seismic actions on structures,

1 9 99.

306
Level 1- Objective (Goals)

Level 2 ' Function Statements

(Furiction Requirements)

Level 3 • • - Perfor^iance Requii^ment

(Operational Requirements)

Deemed-^o-Satisfy Verification

(Acceptable Solutions) Method \

mandatory and non- mandatory documents

Figure 1 Nordic Five Level System

Earthquake Performance Level

(structural or Non-Structural Systems)

Level 1 Level 2 Level 3 Level 4

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307
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DRAFT Y2K PSH MODEL

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312
Recommended Changes to the AASHTO Specifications for the

Seismic Design of Highway Bridges (NCHRP Project 12-49)

by

Ian M. Friedland, P.E.' and Ronald L. Mayes^

ABSTRACT

Under the AASHTO-sponsored National

Cooperative Highway Research Program, the

ATC/MCEER Joint Venture is conducting a

project to develop new specifications for the

seismic design of highway bridges. These new

specifications will be recommended for

incorporation into the AASHTO LRFD Bridge

Design Specifications, most likely in early 200 1

.

They are expected to be performance-based and

will address state-of-the-art aspects of highway

bridge seismic design, including the

representation of seismic hazard, design and

performance criteria, analysis methods, steel and

concrete superstructure and substructure design

and detailing, and foundation design. This paper

presents an overview of some of the major

philosophical and technical changes being

incorporated into the new provisions.

KEYWORDS: analysis; bridge design;

detailing; performance criteria; specifications

1.0 INTRODUCTION

In August of 1998, a joint venture of the Applied

Technology Council (ATC) and the

Multidisciplinary Center for Earthquake

Engineering Research (MCEER) initiated work

on a project to develop the next generation of

seismic design specifications for highway

bridges in the United States. The project is sponsored by the American Association of State Highway and Transportation Officials

(AASHTO) and is being conducted by the

National Cooperative Highway Research

Program (NCHRP) of the Transportation

Research Board. NCHRP Project 12-49,

"Comprehensive Specifications for the Seismic

Design of Bridges," will resuh in the

development of specifications and commentary

which are expected to be incorporated into the

AASHTO LRFD Bridge Design Specifications.

These will be supplemented by a series of design

examples demonstrating the application of key

features of the new specifications.

The target date for completion of the project is February 2001, at which time the recommended

specifications and commentary will be assessed

by the AASHTO Highway Subcommittee on

Bridges and Structures, with a decision

regarding adoption of these provisions to be

made in May of 2001 . During the course of the

project, three drafts of specifications and

commentary will be produced and reviewed,

prior to completion and submission of the final draft to AASHTO.

^ The new specifications are expected to incorporate all current and state-of-the-art practices in bridge seismic design, and will be

performance-based. They will address the

representation of seismic hazard, design and

performance criteria, analysis methods, steel and

concrete superstructure and substructure design

and detailing, foundation design, and soil behavior and properties. The specifications are

also intended to address the differences in seismic hazard, soils, and bridge construction

types found throughout the United States, and

therefore are to be national in scope.

These new specifications will be a marked

departure from the design philosophy, approach,

and requirements currently in use in the United

States. This paper provides a brief overview and

discussion of some of these new approaches and

procedures.

' Applied Technology Council, Washington, D.C. 20004, ifriedland@atcouncil.org

^ Bridge Engineering Consultant, Moraga, California 94556, ronald.mayes@worldnet.att.net

313
2.0 PERFORMANCE CRITERIA

A performance-based design criteria will be

included in the new specifications and

commentary. Specifically, the project will

recommend adoption of a dual- level performance criteria, and two levels of

performance objectives. The current criteria are described in Table 1

.

The recommended performance objective is now

totally performance based and each State determines the desired performance objective for

any particular bridge. This is a change from

current AASHTO definitions of "other,"

"critical" and "essential" bridges for defining

performance levels. The R-Factors are approximate at this time and are provided to indicate their order of magnitude and relativity.

Note that in the final specification, the R-Factors

will be removed from this table.

Tie upper level earthquake considered in these

provisions is designated the Maximum

Considered Earthquake or MCE. In general, the

MCE ground motions have a probability of

exceedance of 3% in 75 years, which is an approximate return period of 2500 years. However, MCE ground motions are bounded

deterministically to values lower than 2500 year

ground motions adjacent to highly active faults.

The current AASHTO specifications have a 500

ft span length limitation. The new specifications

will proposed that this be removed, and a review

conducted to determine what, if any, limitations

should be imposed on the use of the criteria once

its development is more complete.

3 .0 SERVICE AND DAMAGE LEVELS

3.1 Service Levels

Two definitions of service levels have been

included in the first draft of the new

specifications:

Service Level - Immediate (R = 0.67 to 2) - Full access to normal traffic should be expected

following the earthquake; an inspection of the

bridge should be conducted as soon as possible

to ensure that there is no significant damage.

Service Level - Significant Disruption (R = 6 to

8) - Limited access (reduced lanes, light

emergency traffic) may be possible after shoring, however the bridge may eventually

need to be replaced.

An R-Factor of 0.80 is approximately first yield

of reinforcement in a reinforced concrete column whereas an R-Factor of 1 .0 is when the

ultimate moment capacity is achieved and

occurs when all the tensile reinforcement is at yield in a reinforced concrete column or the frill section has reached yield in a steel column.

It is also recommended that all single- and

multi-column bents use the same R-Factor; this

is a significant departure from the current AASHTO provisions which specify different RFactors for single- and multi-column bent.

3.2 Damage Levels

The current draft of the specifications also

contain three damage level definitions:

No Damage - Evidence of movement may be

present but there is no notable damage.

Minimal Damage - Some visible signs of

damage. Minor inelastic response may occur, but post-earthquake damage is limited to narrow

flexural cracking in concrete. Permanent

deformations are not apparent, and any repairs

could be made under non-emergency conditions

with the exception of deck joints. Specific

criteria for geometric and structural constraints are provided.

Significant (Major) Damage - Considerable to

total loss of operations. Although there is no

collapse, permanent offsets may occur and

damage consisting of cracking, reinforcement

yield, and major spalling of concrete may

require closure to repair. Partial or complete

replacement may be required in some cases. Again, specific criteria for geometric and

structural constraints at various service levels are provided.

314
An attempt at providing more specific criteria for concrete columns is also being made, as shown in Table 2. These results are then used in conjunction with a pushover analysis to determine displacement (drift) limit states. In columns (2) to (4) of the table, nominal strain limits are given. From these values normalized

plastic curvatures {^pD) are computed and

presented in column (5). In providing these

values, it is assumed that the code specified

detailing provisions for transverse reinforcement

are used. Then, by assuming the equivalent

(rectangular) plastic hinge length is 45% of the

member diameter/width, it is possible to calculate a nominal plastic hinge rotational

capacity as given in column (6). From this,

dependable (conservative) values are set for code specified rotational limits in column (7). These specified values may be used in lieu of a more precise analysis.

Alternately, a moment-curvature analysis based on strain compatibility and nonlinear stress- strain relations can be used to determine plastic

limit states. From this, a rational analysis is used to establish the rotational capacity of

plastic hinges.

4.0 GEOMETRIC AND STRUCTURAL

CONSTRAINTS

In the initial phases of this project, an attempt was made to develop a set of geotechnical

performance objectives that would be similar to those being developed for concrete columns. A

two-day workshop was held to review initial draft proposals and to refine recommended

criteria and approaches. The consensus of the

workshop was that the amount of acceptable

foundation and abutment movement should be

related to geometric and structural constraints by

bridge type, rather than explicit values on

foundation movements. As a result, the first draft specification proposes constraints that would implicitly provide foundation design

limits for seismic loads to meet the various

performance objectives. Since this is the first time an attempt has been made at developing

these constraints, preliminary specified values

have been considered, but should only be

considered as approximate or order of

magnitude. Additional work will be required to refine and finalize these constraints.

Geometric constraints generally relate to the

usability of the bridge by traffic passing on or under it. Therefore, such constraints will usually

apply to permanent displacements that occur as a

result of the earthquake. The ability to repair (or

the desire not to be required to repair) such

displacements should be considered when

establishing displacement capacities. When

immediate service is desired, permanent

displacements should be small or non-existent,

and should be at levels that are within accepted

tolerances for normal highway operations.

When limited service is acceptable, the

geometric constraints may be relaxed. These

may be governed by the geometry or types of

vehicles that will be using the bridge after an

earthquake, and by the ability of these vehicles

to pass through the geometric obstruction.

Alternately, a jurisdiction may simply wish to

limit displacements to a multiple of those

allowed for immediate service. In the case of a

significant disruption, post-earthquake use of the

bridge is not guaranteed and therefore no

geometric constraints would be required to achieve these goals. However, because life safety is at the heart of this performance level,

jurisdictions should consider establishing some

geometric displacement limits.

Structural constraints on displacements can be

based on the requirements of a number of

structural elements and can result from either

transient displacements due to ground shaking or permanent displacements resulting from ground

movement due to faulting, seismically induced

settlements, lateral spreading, and so forth. Structural damage to foundation elements is limited primarily to piles since footings and pile

caps are usually capacity protected. Although

pile damage can often be avoided at a reasonable

cost when piles are capacity protected, requiring

this for all piles could lead to overly

conservative, and thus expensive, foundation

designs. Therefore, it may be desirable to

establish some lateral displacement limits for

piles based on a limited amount of structural

315
damage that is unlikely to compromise the

structural integrity of the bridge. These limits

will be based on the type and size of the piling,

and the nature of the soil near the head of the

pile. Caltrans has attempted to establish such

limits for its standard piling based on physical

testing. Other jurisdictions may attempt to do

the same, or they may perform more complex

analytical studies to establish similar

displacement, capacity, and stiffness limits.

5 .0 DESIGN-EVENT RETURN PERIODS

Current AASHTO specifications considered a

single-level earthquake hazard design event,

based on a 500-year return period. At the time

this event was incorporated into the AASHTO

specifications, it was the only event readily

available as a design value for seismic hazard

representation in the United States.

For this project, it has been recommended that a 2,500-year return period, which provides an

equivalent 3% probability of exceedance in 75

years, be used as the upper level design event. A

lower level event with an approximate 100-year

return period, which would be based on a 50%

probability of exceedence in 75 years, has also

been recommended. In brief, the reasons for these recommended return periods include:

• For a frequent or expected earthquake (50%

probability of exceedance in 75 years), this

is the design level under which the bridge

should remain essentially elastic (R = 1 .0). This will result in a performance that is equivalent to an elastic (no damage) design

for a 100 year flood, and is less conservative

but similar to an elastic design for a 100

mph base wind design.

• For a rare earthquake (3% probability of

exceedance in 75 years), this design level is recommended in order to assure that a "no

collapse" performance criteria for a Maximum Considered Earthquake (MCE) is satisfied. Since seismic loads increase so

significantly compared to wind and flood

loads as the return period increases,

earthquake design needs to consider longer

return period design events.

A critical earthquake design issue in preventing

collapse of a bridge is to ensure that deck

displacements and hence seat widths for the

girders can accommodate displacements from

events that have occurred and can be realistically

expected to reoccur. A return period of 500

years does not come close to capturing the force

and displacement levels that may have occurred

during earthquakes in the New Madrid region

(181 1-1812) and Charleston, South Carolina

(1886). A return period of the order of 2500

years is required to obtain realistic displacement

levels that may have occurred during these

earthquakes.

It is noted that the MCE ground motion maps

incorporate deterministic bounds on the ground

motions adjacent to highly active faults. These

bounds are currently applicable in California,

along a portion of the California-Nevada border,

in coastal Oregon and Washington, and in parts

of Alaska and Hawaii. These deterministic

bounds are applied so that the ground motions

do not become unreasonably large in comparison to the ground motions that could be

produced by maximum magnitude earthquake on

the faults. Deterministic bounds are defined as 150% of the median estimates of ground

motions calculated using appropriate attenuation

relationships, and assuming the occurrence of a maximum magnitude earthquake on the fault. However, they are limited to not less than 1 .5 g

for the short-period spectral acceleration plateau

and 0.6g for 1.0-second spectral accelerations.

6.0 SHAPE AND CONSTRUCTION OF

THE DESIGN RESPONSE SPECTRA

The long period portion of the current AASHTO

acceleration response spectra is governed by the

spectrum shape and the soil factor, and it decays

as 1/T2/3. There was considerable "massaging"

of the factors that affect the long period portion

of the current AASHTO spectra in order to produce a level of approximately 50%

conservatism in the design spectra when

compared to the ground motion response spectra

beyond 1 second period. The reasons given for

^ this level of conservatism included:

316
(a). The fundamental period of a bridge

increases as (1) the column height

increases, (2) the span length increases,

and (3) the number of columns per bent

decreases. Hence, the longer the period,

the more likely that high ductility

requirements will be concentrated in a few columns.

(b) Instability of a bridge is more of a problem as the period increases.

Analysis of ground motion data indicates that

the acceleration spectra generally decreases with

period in the long period range as 1/T or more

rapidly. The shape of the long period portion of

the recommended response spectra is less conservative than the current spectral shape and

it is recommended that the shape of the spectra

decay as 1/T.

For periods longer than about 3 seconds,

depending on the seismic environment, use of

the 1/T relationship for spectral acceleration

may be conservative because the ground

motions may be approaching the constant

spectral displacement region for which spectral

accelerations decay as 1/T2. Either the 1/T

relationship may be conservatively used or a

site-specific study conducted to determine an

appropriate long-period spectral decay.

The U.S. Geological Survey (USGS) has

prepared MCE ground motion maps for spectral

accelerations at 0.2 and 1 .0 second periods. In

addition, the 0.2-second and 1 .0 second spectral

accelerations can be obtained at other

probability levels from a CD-ROM published by

the USGS. Response spectra can be constructed

using the spectral accelerations at these two

periods and the spectrum construction method

illustrated in the following figure. This

spectrum construction is denoted herein as the

"two-point" method because it involves

intersecting the constant spectral acceleration

plateau passing through 0.2 second, with the

long-period branch of the spectrum that declines

as 1/T and passes through 1.0-second spectral

acceleration. This is also the spectrum

construction method that has been adopted in the

1997 NEHRP Provisions, will continue in the

2000 NEHRP Provisions, and is proposed for

the 2000 International Building Code.

7.0 DESIGN PROCEDURES AND

RESPONSE-MODIFICATION

FACTORS

The current draft of the specifications provide

for three levels of design and analysis

procedures in the moderate to high-seismic

zones. This is in addition to the current no

seismic or low-level minimum requirements,

which are defined as Level 0:

Level 0 - These are the current AASHTO

Zone 1 provisions, which require

minimum seat widths and specified

design forces for fixed bearings of

10% dead load for Zone lA and

25% dead load for Zone IB.

Level 1 - These requires no formal seismic

analysis but requires the use of

capacity design principles and

minimum design details.

Level 2A - This is a one-step design procedure

based on a new analysis method

referred to as the "capacity spectrum

method" and is applicable to very

regular bridges. This method has

been incorporated in some of the

retrofit guidelines for buildings and

provides the designer with the

ability to assess whether or not a

bridge designed for all non-seismic

loads has sufficient strength and

displacement capacity to resist the

seismic loads.

Level 2B - This is a one-step design procedure

based on an elastic (cracked section

properties) analysis using either the

Uniform Load or Multimode

method of analysis. The analysis is performed for the governing design

spectra (either 100 year or 2500 year

event) and the use of a conservative

response modification factor (Rfactor) (i.e. 5 rather than 8 for 2500

year event for columns). The

317
analysis will determine the moment

demand at all plastic hinge locations

in the column. Capacity design

principles govern foundation and

column shear design. If sacrificial elements are part of the design (i.e. shear keys) they must be sized to

resist the 1 00 year forces and the

bridge must be capable of resisting

the 2500 year forces without the

sacrificial elements (i.e. two

analyses are required if sacrificial elements exist in a bridge).

Level 3 - This is a two-step design procedure

using an elastic (cracked section

properties) analysis with the

multimode method of analysis for the governing design spectra ( 1 00

year/1 .0 or 2500 year/8) to perform

preliminary sizing of the moment

capacity of the columns. Capacity

design principles govern foundation

and column shear design. A

pushover analysis is then performed.

The designer can change the design

forces in the columns provided they

are not lowered below those

required by the 1 00 year event and

they satisfy the displacement

demand.

The attempt in this project is to make the Level

1 approach applicable to as wide a range of

bridges as possible. Level 2A will be applicable

to very regular bridges that can be idealized as Single-Degree-of-Freedom systems, until more

development woik is done to extend the method.

It is also recommended that for Levels 2B and 3, a quick assessment as to which of the two

earthquake levels govern column flexural

moment capacity is quick and easy to perform

and is worthwhile.

The recommendation is to have both a one-step,

two-level (Level 2) design/analysis procedure,

and a two-step, two-level design/analysis

procedure (Level 3). The one-step Level 2B

procedure would be consistent with the current design requirements but would have more

conservative R-factors than the two-step

procedure (e.g. 4 vs. 6 or 5 vs. 8). It also would

use the more conservative design values

resulting from the 100 year or 2500 year events. The Level 2B procedure, which is the same as current AASHTO requirements, in summary

would be as follows:

a) Perform an analysis using the

appropriate 100 year or 2500 year

design spectra. Two analyses may be

necessary if sacrificial elements such as shear keys are part of the design.

b) Divide by specified (conservative) Rfactors and design the components using

capacity design principles.

The Level 3 procedure would include a) and b)

but the R-Factors would be approximately 150%

higher than the one-step (Level 2) procedure.

The second essential step in Level 3 is to perform a "pushover analysis." The objective of

the pushover analysis would be to ensure that

the maximum plastic rotation and drift limits for

the various materials and components are

satisfied, and that the displacement capacity of

each column exceeds the displacement demand

by a factor of approximately 1.5. It is anticipated that an upper and lower bound

pushover analysis would be performed using

upper and lower bound soil properties.

Furthermore a designer could change column

design forces as part of this process provided

they were never less than those required by the

1 00 year event.

Since the response spectra for the various return periods (2500 year and 100 year) are quickly

and easily available from the USGS on the

Internet, it will be relatively easy to determine

which one would give the highest forces for

determining the column flexural strength

It is also proposed to use the same R-factors for

both single and multi-column bents. This is contrary to the current AASHTO specifications

and arises from studies performed during the

development of recommendations for Caltrans.

Initial research conducted for Caltrans actually

recommended higher R-factors for singlecolumn bents than for multi-column bents,

318
which is the opposite of current AASHTO

requirements. These recommendations arose out of the relative displacement ductilities that could

be achieved from each system, given constant curvature ductilities in the plastic hinge zones. This is true because the displacement at first yield in a multi-column system with "pinned"

column bases comprises a larger percentage of

the ultimate displacement than does first yield

displacement in a single column system.

Therefore, the single column system should be more ductile.

The order of magnitude of R-factors being

considered is given in Table 3. These will be

further refined and evaluated in subsequent work

on the project. The numbers provided show the

relativity between the one-step (Level 2) and

two-step (Level 3) procedures and between the

critical and regular bridges. It should be noted

that the impact of the R-ractors on the design of

a column is not as important as the minimum

longitudinal steel requirements and the P-A

limits, and that R-factors higher than 6 virtually never govern. It is therefore likely that the Rfactor for single and multi-column bents will not exceed 6.

8.0 LEVEL 1, NO SEISMIC ANALYSIS

DESIGN REQUIREMENTS

The "no seismic analysis" design procedures are an extremely important part of the

recommendations because they will apply as widely as possible in Zone 2. The purpose of

the provisions is to provide the bridge designer

the ability to design structures without the need

to undertake dynamic analyses. The bridge

elements are designed in accordance with the

usual provisions, but the primary seismic

resisting elements are detailed for seismic

resistance.

There are capacity design requirements for shear

reinforcement and confining reinforcement at the plastic hinge locations in columns. There are also capacity design requirements for detailing

the connection to pile and column caps. There

are no additional design requirements for the

abutments or foundations with one exception -

integral abutments need to be designed for

passive pressure,. The premise for this is based on the parameter study which demonstrated that a foundation designed using a factor of safety of

3 will perform well in these lower zones assuming the column connection to the pile cap

is based on capacity design principles. These

procedures will also be permitted for use in sites with liquefaction potential, will but require some additional considerations if the

predominant moment magnitude, Mb , for the

site exceeds 6.

The limitations on the applicability of the

provisions initiated with the definition of a

"regular bridge" as contained in the current AASHTO seismic design provisions. Additional

limits were added based on engineering

judgment, and these need to be evaluated in future studies. The limits as they currently exist

consider (a) regularity, (b) axial load limits, (c)

load path and force sharing, and (d) continuity.

Table 4 provides the current definition of a

"regular" bridge.

Span-to-span continuity must be provided by one

or more of the following means:

• A superstructure that is monolithic with

the substructure

• A superstructure seated on bearings that has transverse restraint

• Simply supported girders seated on

bearings must be fixed at one support

end, and transversely restrained at the

other

• A superstructure supported on isolation

bearings that act in all directions that accommodate displacements

9.0 OTHER ISSUES

9.1 Foundations

Provisions for spread footings, driven piles,

drilled shafts (or cast-in-drilled-hole piles), and

abutments are being significantly improved over

similar provisions in current AASHTO

specifications. These are, in large part, based on

a number of advances made through research

programs sponsored recently by the Federal

319
Highway Administration, Caltrans, and others. For abutments, explicit recognition of sacrificial elements is being provided, including knock-off

back walls and other similar fuse elements.

9.2 Steel Superstructures

The current AASHTO specifications have

minimal requirements for steel bridges. As a

result, a substantial effort is being made in this project to develop appropriate provisions for

steel bridges. A first draft of such provisions

has been prepared, and will be reviewed during a workshop to be held in the summer of 2000 by

an invited select group of consulting engineers,

transportation agency staff, and researchers, all experts in seismic design and steel bridges. It is anticipated that this workshop will provide the

necessary framework to resolve outstanding

issues, and that a consensus will be reached

regarding the appropriate scope of provisions for the seismic design of steel bridges.

9.3 Special Considerations Regarding

Seismic and Geotechnical Hazards

A number of special provisions are being

included to address important considerations in seismic hazard. Included in these are specifications related to vertical accelerations

and their impacts on bridge performance, and

near fault (and so-called "fling") effects. A

significant amount of work is also being done to address the issues of soil liquefaction and lateral spreading, especially as this relates to the

infrequent but potentially large upper-level

2500-year event.

10. CONCLUSIONS

The new specifications for the seismic design of

highway bridges being developed under NCHRP

Project 12-49 will be a significant departure

from the current procedures and requirements

found in AASHTO specifications. Work on NCHRP Project 12-49 is expected to be

completed in the spring of 200 1

, and the

AASHTO Highway Subcommittee on Bridges

and Structures will then review the provisions

resulting from the project, with possible

adoption by AASHTO in late 2001.

320
Table 1 - Performance Objectives

Performance Objective

Probability of Exceedance

(Approximate Return Period) Life Safety Operational

For Design Ground Motions

Rare Earthquake Service Significant Disruption Immediate

3% in 75 years R Factor R = 6to8 R = 2

(approx. 2500 Year) Damage Significant Minimal

Frequent Earthquake Service Immediate Immediate

50% in 75 Years R Factor R^LO R = 0.8

(approx. 100 Year) Damage Minimal None

Table 2 - Example of Qualitative Specified Column Performance Requirements

Service Level

(1)

Concrete Strains

Steel

Strain

(4)

Plastic

Curvature

(5)

Plastic Hinge Rotations

0p (rad)

Cover

(2)

Core

(3)

Nominal

Capacity

(6)

Dependable

(Code)

Capacity

(7)

Fully functional .003 0.002 0.002

Life safety spalled 0.045 0.045' 0.09' 0.04' 0.03'

May be governed by low cycle fatigue of the longitudinal reinforcement

Table 3 - Typical Base Response Modification Factors

Substructure Element Importance Category

Other Essential Critical One Step

Design

Two Step

Design

One Step

Design

Two Step

Design

Two Step

Design

Wall Piers-larger dimension 2 3 1.5 2 1.5 Columns - Single and Multiple Ductile

Detailing

4 to 5 6 to 8 2 3 2

Columns - Singe and Multiple Minimum

Detailing (see Comments below)

3 4 1.5 2 1.5

Pile Bents and Drilled Shafts - Vertical

Piles - above ground

4 6 2 3 2

Pile Bents and Drilled Shafts - in ground 2 3 1.0 1.5 1.0 Pile Bents - Batter Piles 3 5 1.5 2 1.5 All Elements - 150 year EQ NA 1.5

(1 for wall piers)

NA 1 .75

321
Table 4 - Parameter Definitions for "Regular" Bridges

Parameter Value

Number of Spans 2 3 4 5 6

Maximum subtended

45 45 45 45 45

angle for curved bridge

Maximum span length

ratio, from span to span

3 /. 2 1.5 1.5 Maximum bent/pier

stiffness ratio from span 4 4 3 2

to span excluding abutments

322
To = 0.2Ts

Fa X Sar(0.2)

11 Sar(0.

M

Fv X Sai

^ " Fa X Sai

r(1.0)

'(0.2) N(ac

TV

9te: Sar is i ;ceieration rpe B

spectral on rock

\ /— V X oar(i .u T

rSar(1.or Soil

x-spe<

response

Jtrum

Sa=0.4Fa X Sar(0.2) 5ar(1.0)^

T

1

1

Rock (Type

responsespectnim

(0 CO

c

COha

<»o

u

<

CO tsoa

(/>

OToO^ Ts

Period, T (sec)

Figure 1 - Response Spectrum Construction

323

EARTHQUAKE RESISTANT DESIGN OF PORT FACILITIES

by

Takashi NAGAO", Atsushi NOZU^^

ABSTRACT

This paper presents the earthquake resistant design way of port facilities in Japan. Technical

Standard of Port and Harbour Facilities was first established in 1973, and has been revised 3

times so far. Having experienced the 1995

Hyogoken-Nambu earthquake, significant

revision has been made on the outcome of the recent research.

KEY WORDS : Earthquake Resistant Design

Port Facility

Design Code

1. HISTRY OF REVISIONS OF DESIGN

CODES

Having been established in 1951, the Port and

Harbour Law in Japan has been revised many

times so far. The important revision in view of

the design of port and harbour facilities was made in 1974, in which it was noticed that the

port and harbour facilities must be constructed,

maintained and rehabilitated in accordance with

the Technical Standard of Port and Harbour

Facilities. In 1975 the engineering requirement was established as the Ordinance of the Ministry

of Transport and it was prescribed in the

ordinance that the facilities in ports and harbours

must be stable against the loads such as earthquake loads, dead weights, wave forces,

impacts due to ships and so on. The Technical Standard of Port and Harbour

Facilities was established in 1973 as the order of

the Director General of Bureau of the Ports and

Harbours, Ministry of Transport, in which the

details on earthquake resistant design, such as design procedures, factor of safety and

allowable stresses, were specified. In 1979 the Technical Standard of Port and

Harbour Facilities and its Commentary was

compiled under the supervision of the Bureau of

the Ports and Harbours, Ministry of Transport,

and has been revised in every ten years after the

1979 edition. Seismic stability of the port and harbour

structures was to be examined only by the

pseudo-static method in the 1979 edition and

1989 edition of the Technical Standards. The

pseudo-static method is called the seismic

coefficient method, and the earthquake load is obtained by the multiplication of the design

seismic coefficient and the vertical load. The

design seismic coefficient is obtained by the

multiplication of the regional seismic coefficient,

the factor for subsoil condition, and the

importance factor. Those three factors were

classified into three groups respectively, with the

regional seismic coefficient ranging from 0.05 to 0.15, the factor for subsoil condition ranging

from 0.8 to 1.2, and the importance factor ranging from 0.5 to 1.5. The resultant value, the

design seismic coefficient, was rounded off to the nearest 0.05 or 0.00. As to the design of the

reinforced concrete structures, allowable sfress method was applied.

1) Chief, Geotechnical Earthquake Engineering

Laboratory, Structural Engineering Division, Port and Harbour Research Institute,

Ministry of Transport

2) Senior Research Engineer, Structural Engineering Division, Port and Harbour

Research Institute, Ministry of Transport

325
Significant modification had not been made as

to the earthquake resistant design procedures

ever since the first edition thus far, however, the

procedure of assessing the liquefaction potential was not stated in the 1973 edition, and was

firstly stated in the 1 979 edition. In 1999, the order of the Director General of

Bureau of the Ports and Harbours was repealed

for variety of reasons, and the Ministry of

Transport notified the new detailed Technical

Standard. In the new Technical Standard, some

significant revisions have been made based on

the outcome of the recent research after the 1 995

Hyogoken-Nambu earthquake. Those are summarized as follows :

( 1 ) Principles of design

The concept of perfoiTnance-based design

has been introduced. The principles are: Q All the structures must be stable against

the level 1 earthquake motions whose

return periods are about 75 years. ©High seismic resistant facilities should

keep the required performance against

the level 2 earthquake motions whose

return periods are over some hundred years.

(2) Seismic coefficient method

Q The regional seismic coefficient and

the importance factor have been modified,

while the factor for subsoil condition has

remained as it was. The number of regional

groups for the regional seismic coefficient

has come up to five, coefficient ranging

Table.l Summary of history

from 0.08 to 0.15. Range of the importance

factor has become from 0.8 to 1.5. In

addition, the resuhant value has been

considered down to three decimal places.

(2) The equation for the apparent seismic

coefficient, which is the seismic coefficient

used for the calculation of earth pressure below groundwater level, has been

modified.

(3) Consideration of the dynamic water

pressures acting at the front of vertical walls has been stated.

(3) Assess of earthquake-resistant performance

Assessing way of the earthquake-resistant

performance in view of level 2 earthquake

motions has been introduced.

(4) Assess of liquefaction potential

Assessing way of liquefaction potential has

been modified.

(5) Design method of open piled piers

Modified pseudo-static design method,

which is called the modified seismic

coefficient method, has been introduced for the design of open piled piers.

(6) Design method of reinforced concrete

structures Limit state design method has been

introduced, and safety factors for the design

have been established.

The history of revisions of design codes is summarized in Table 1

.

of revisions of design codes

1973 1979 1989 1999

Earthquake design level One level Two levels

Seismic coefficient method o

o

modified

Performance-based design principles o

Assess of liquefaction potential o

o

modified

Design of open piled piers

Seismic coefficient

method

Modified seismic coefficient

method

Design of reinforced concrete structure Allowable stress method Limit state design method

326
2. SUMMARY OF THE UP-TO-DATE

DESIGN CODE

(1) Evaluation of Seismic Performance

(a) General

In the design of port facilities, the effect of

earthquakes should be taken into account so

that they possess appropriate amount of

seismic resistance.

Explanation

At Kobe Port, the type of structures were quite

uniform during the 1995 Hyogoken-Nanbu

earthquake. This is why ahnost all of the

structures suffered similar damage. If the type of

structures had been more diverse, the amount of

damage for each structtu-e should not have been

unifomi because their response characteristics should have been different. Based on this experience, in the choice of the structural type of

port facilities, it is recommended to adopt

various type of structures as long as possible.

(b) Seismic performance requirement for

port facilities.

(1) Port facilities should sustain their structural stabiUty and maintain their functions for a

level- 1 groimd motion, which by definition occurs with high probability during the

facility's duration.

(2) High seismic resistant facilities, which are

especially important and require high

seismic resistance, are allowed to suffer only slight damage for a level-2 ground

motion, which by definition occurs with

relatively low probability during the

faciUty's duration but which is very intense. In other words, high seismic resistant faciUties should be prepared for rapid

restoration to sustain their intended

functions after a level-2 ground motion.

Explanation

In the seismic design of port structures, a level 1 ground motion, which has a return period of 75

years and a level 2 ground motion, which is a ground motion due to intra-plate earthquake

with a return period of more than several hundred years or a ground motion due to a subduction zone earthquake, should be taken

into account. High seismic resistant facilities include high

seismic resistant quay walls, which are specially

designed for the transportation of emergency

cargo or for the maintenance of economic or

social activity just after the earthquake, and the revetments of the disaster prevention base,

which is intended to keep the safety of the

citizen just after the earthquake.

While 'to maintain their functions' means to sustain their structural stability, 'to sustain

intended functions' means to suffer only a slight

damage and to be prepared for a immediate

restoration.

(c) Pseudo-static method

( 1 ) In principle, seismic load for port structures with relatively short natural period and

relatively high damping factor should be

designated as a design seismic coefficient for pseudo-static approach. In this case, the

design seismic coefficient should be used.

Seismic inertia force should be the larger of

the following (a) and (b) and should be

assimied to act on the gravity center of the

structure.

(a) (Seismic force)=

(Self weight) X (Seismic coefficient)

(b) (Seismic force)=(Self weight +

Surcharge) X (Seismic coefficient)

(2) For structures for which pseudo-static

method is not applied, seismic load should

be designated in an appropriate maimer,

taking into account the characteristics of

the structure.

327
Explanation

(1) For quay walls and other similar port

structures, pseudo-static method is applied

as for other wide range of structures. Because natural periods of these structures are generally shorter than predominant

periods of ground motions, the response of

these structures during earthquake are

similar to those of rigid bodies on a rigid

table.

(2) Because the seismic load is assumed to act as a static load in the pseudo-static method,

it is necessary to take into account the

difference between the real phenomena and

the assumptions in the method. To appreciate

this difference, the safety factor and the

allowable stress for dynamic loads are

different from those for static load.

(3) It is preferable to examine the seismic

resistance of those structures which has a longer natural period compared to predominant periods of ground motion or for which the distribution of acceleration is not uniform along the height.

(d) Earthquake response analysis

If the facility is especially important or the

type of structure is rare and there is no similar

conventional structure, it is recommended

that it's seismic resistance should be

examined by using earthquake response

analysis together with conventional pseudostatic method or seismic deformation method.

Explanation

Recently, new type of port facilities or extremely large port facilities have been

designed and constructed. On the other hand, it

is sometimes required to construct port

structures at a site with a poor subsoil conditions.

Furthermore, it is requested to examine whether

a high seismic resistant facility will maintain

their functions after a near-source ground

motion such as the ground motion at Kobe Port during the 1 995 Hyogoken-Nanbu earthquake. It

is recommended to examine earthquake

resistance of structures by conducting

earthquake response analysis to understand the performance of structures during earthquake more precisely if the type of structure is new or

if the structure is especially important.

(e) Seismic deformation method

Because the deformation of line structures etc. buried in the soil during earthquake is controlled by surrounding soil, it is preferable

that such structures should be designed by

using seismic deformation method.

328
(2) Earthquake Load

(a) Design Seismic coefficient

(1) For pseudo-static design of port structures, horizontal design seismic coefficient should be

determined with following equation.

Seismic coefficient =Regional seismic coefficient X Factor for subsoil condition X Importance factor

(1)

Horizontal design seismic coefficient should be rounded to obtain two places of decimals. Standard

values for regional seismic coefficient are: Region A: 0.15

Nemuro, Kushiro, Tokachi and Hidaka districts of Hokkaido, Saitama, Chiba, Tokyo (Except for Hachijo and Ogasawara Islands), Kanagawa, Yamanashi, Shizuoka, Aichi, Gifu, Fukui, Shiga, Mie,

Nara, Wakayama, Osaka and Hyogo.

Region B: 0.13

Pacific side of Aomori, Iwate, Miyagi, Fukushima, Ibaragi, Tochigi, Gunma, Nagano, Kyoto, Kochi

and Tokushima.

Region C: 0.12

Iburi, Oshima and Hiyama districts of Hokkaido, Aomori (except for Pacific side), Akita, Yamagata,

Niigata, Toyama, Ishikawa, Tottori, Hiroshima, Ehime, Oita, Miyazaki, Amami Islands of Kagoshima

and Kumamoto.

Region D: 0.11 Abashiri, Goshi, Ishikari, Sorachi, Rumoi and Kamikawa districts of Hokkaido, Okayama, Tottori,

Kagawa, Nagasaki (except for Goto, Dei and Tsushima Islands), Saga, Kagoshima (except for Amami

Islands) and Okinawa (except for Daito Islands).

Region £: 0.08

Sorachi district of Hokkaido, Hachijo and Ogasawara Islands of Tokyo, Yamaguchi, Fukuoka, Goto, Iki and Tsushima Islands of Nagasaki and Daito Islands of Okinawa.

Factor for subsoil condition should be determined as shown in Table 2 and 3.

Table 2 Factor for subsoil condition

Classification 1^ kind 2"^ kind 3"* kind

Factor 0.8 1.0 1.2

Table 3 Classification of subsoil

Thickness of

Quaternary

Deposit

Sand

Gravel or

clay

Soft ground

less than 5 m

5-25m

more than 25m

JS. JS.

^rd

2nd

^rd yd

329
Importance factor should be determined according to Table 4.

Table 4 Importance factor

Category Special A B c

Factor 1.5 1.2 1.0 0.8

Category Special : The structure has significant characteristics described by items (l)-(4) of category

A.

Category A: (1) If the structure is damaged by an earthquake, a large number of human life and

property will possibly be lost. (2) If the structure is damaged by an earthquake, economic or social

activity of the region will be severely suffered. (3) The structure will perform an important role in the

reconstruction work of the region after the earthquake. (4) The structure handles a hazardous or a dangerous object and it is anticipated that the damage of the structure will cause a great loss of human

life or property. (5) If the structure is damaged, it is supposed that the repair work is considerably

difficuh.

Category B: The structiire does not belong to categories Special, A nor C.

Category C: The structure does not belong category Special nor A and is easy to repair or, even if the

structure is damaged by an earthquake, the effect on economic or social activity is small.

(2) If vertical seismic coefficient is required in the pseudo-static design, the vertical seismic coefficient

should be determined appropriately, taking into account the characteristics of structure and subsoil.

Explanation

(1) To determine importance factor of the

structure, it is necessary to consider not only

the purpose, type or size of the structure but

also social or economic aspects of the

structure. Following factors also should be

taken into account.

a) The extent of damage in the future

earthquake, the difficulty of restoration work

or the residual strength after the earthquake.

b) The cargo-handling capacity of the

other facilities of the same port.

Therefore, it is possible to use different

importance factors for the structures for the

same cargo in the same port when desired.

(2) For structures other than high seismic

resistant quay walls, the upper limit of design

seismic coefficient should be 0.25 for several

reasons. First, in the past, the upper limit of

design seismic coefficient was 0.25. Second,

there has been no port structures with design

seismic coefficient of 0.25 that suffered

significant damage. Thirdly, high seismic

resistant quay wall has been constructed in

many ports.

330
(3) Seismic Design of High Seismic Resistant

Quaywalis

(a) Evaluation of seismic performance of higli

seismic resistant facilities

(1) In the design process of high seismic

resistant facilities, it is requested that

their seismic performance should be

evaluated for a level-2 ground motion to assure that their seismic resistance is

satisfactory.

(2) Seismic performance should be evaluated

by appropriately modeling the soil and

the structure of the facility, with a method

which is appropriate for the particular

type of the structure.

Ground motion which is used for the evaluation

of seismic performance should be determined

with response analysis of the ground in principle.

Explanation

(1) Evaluation of the residual deformation of

high seismic resistant facilities, which is based on a earthquake response analysis, is

required for the purpose of verifying that they

will sustain their intended functions after a

level-2 ground motion. The reason is that, for

the examination of the stability of the

structure or the soil for a large ground motion

such as a level-2 ground motion,

conventional pseudo-static method is not

sufficient.

(2) The judgement whether the high seismic

resistant facilities will sustain their intended

functions based on the results of earthquake

response analysis should be based on the

combined considerations on the stability of

the structure after the earthquake, the

ftinctions and the difficulty of restoration

work. Although the allowable residual

deformation should be defined for this

judgement, it is not easy to specify the

allowable deformation at the present state of

knowledge. In the case of the 1995

Hyogoken-Nanbu earthquake, some of the

caisson quaywalis with a normalized

deformation (lateral residual displacement /

height of the quaywall) of over 10-20% was

temporary repaired and offered for immediate

use just after the earthquake.

(b) Design Seismic coefficient of high seismic

resistance facilities

(1) When pseudo-static design is apphed to high seismic resistant quaywalis, the design

seismic coefficient should be determined by

a global judgement base on the seismic

coefficient determined by Eq.(l) with

importance factor 1.5, by following

equations for which peak ground

acceleration should be calculated with

ground response analysis for level-2 ground

motion, and by other appropriate methods.

1 . If a is smaller than or equal to 200Gal,

Kh=a/g (2)

2. If a is larger than 200Gal,

K,=(l/3)X(a/gy"'' (3)

Here, Kh is horizontal seismic coefficient, a is peak ground acceleration at free surface and g

is the acceleration of gravity.

Explanations

(1) When the design seismic coefficient can be

accurately determined by investigating

regional seismic activity, characteristics of

ground motion, site response, etc., it is

preferable to use this design seismic

coefficient instead of the value designated

here. For example, when the design ground

motion is determined based on the

information regarding regional seismic

activities or based on strong ground motion

observations or when seismic response

analysis of the structure is conducted, design

seismic coefficient can be determined based on these results.

331
(2) In the design of high seismic resistant

faciUties, target earthquake has to be selected

from earthquakes including hypothetical

earthquake in the disaster prevention plan set by local government.

(3) One way of calculating peak ground

acceleration at free surface is to use multiple

reflection model for the response analysis of

the ground.

(4) From the experience of significant damage at Kobe Port during the 1995 Hyogoken-Nanbu

earthquake, minimum design seismic

coefficient for high seismic resistant facilities should be 0.25 if the site is in a near-source

region.

(5) When it is desired, seismic resistant qua walls

should be designed for level-2 ground motion

with a method other than pseudo-static

method such as earthquake response analysis. In this case, it is necessary to make sure that seismic resistant facilities will sustain their

structural stability for level- 1 ground motion.

332
NIST Research on Structural Performance of Housing Systems

by

Fahim Sadek' and Michael A. Riley"

ABSTRACT

An outline of NIST research in the area of

structural performance of housing systems is presented in this paper. The performance of

wood-frame houses in past earthquakes and hurricanes is summarized, and the state-of-the-art in experimental and analytical research conducted on wood-frame components and systems is pre- sented. While significant progress has been

made in analytical modeling and testing of shear

walls and diaphragms, only limited progress has

been made on the testing and modeling of complete houses and inter-component connections.

Ongoing and future research efforts at NIST are introduced, and the program objective and technical approach are presented. Finally, a brief

discussion of current experimental work, along

with some initial results, is presented and dis- cussed.

KEYWORDS: Housing systems; Inter- component connections. Nailed connections;

Roof trusses; Shear walls; Wood-frame construction.

1 INTRODUCTION

The majority of residential construction in the

United States is wood-frame buildings. These

buildings perform well under gravity loads, but

considerable damage has been observed in such

structures after significant earthquakes and major hurricanes. To enhance the resistance of

houses to natural disasters and to reduce the risk to life and property, the behavior of wood-frame

buildings subjected to such loads needs to be

understood.

Wood-frame buildings are constructed from several diaphragms such as walls, floors, and roofs; joined by inter-component connections such as

nails, anchor bolts, metal plates, and other propriety connectors. The performance of these

buildings is influenced by the behavior of their individual components and connections. Therefore, an understanding of the behavior of the

different structural components and connections

is essential to accurately predict the performance

of a housing unit under different types of loading.

Structural performance parameters can be relia- bly predicted by using accurate analytical tools

to model the behavior of a complete structure. Multiple simulations can be used to study the

influence of different parameters on the structural performance, and they will be more cost- effective than experimental testing. Experiments, however, are still required to validate and

refme the analytical tools. The result promises

improved design standards that enable construction of buildings that resist earthquake and wind

loads without significant damage or collapse.

Analysis of the structural response of woodframe building components subjected to lateral loads is a difficult task due to several sources of

nonlinearity, the complex nature of the connections, and the wide variability in material properties and construction techniques. Since shear

walls are the most important elements for re- sisting lateral loads, they have been experimentally and analytically studied since the early

1970s. Simplified methods of wall analysis and

finite element models have been developed to predict the behavior of the walls under static and

dynamic lateral loading. The development of

tools for analyzing complete wood-frame houses

' Research Associate, Southern Methodist University, Dallas, TX 75275; on Assignment, National Institute of Stan- dards and Technology, 100 Bureau Dr., Stop 86 11, Gaithersburg, MD 20899-861

1

^ Research Structural Engineer, National Institute of Standards and Technology, 100 Bureau Dr., Stop 8611,

Gaithersburg, MD 20899-861

1

333
started in the mid-eighties. Few investigators

attempted to analyze the different structural components and the inter-component connections in an assembled model.

The scope of NIST research will include developing experimentally validated, three- dimensional, analytical models to predict the

performance of housing units during natural disasters. This model will be used to establish per- formance characteristics of conventional and

innovative housing systems. Testing will be

conducted on components for which modeling

and performance data are lacking, such as inter- component connections.

The following sections summarize the structural performance of wood-frame houses in past

earthquakes and hurricanes, and present the

state-of-the-art in experimental and analytical

research conducted on wood-frame components

and systems. A summary of the NIST research

is presented, followed by a description of ongoing experimental work on the performance of

inter-component connections.

2 PERFORMANCE OF HOUSES IN PAST

EARTHQUAKES AND HURRICANES

Although recent earthquakes and hurricanes in the United States have resulted in few casualties,

they have caused substantial damage, significant

economic loss, and disruption of social and

commercial activities. The performance of resi- dential housing during such events has shown

that wood-frame construction generally meets

the life safety objective. Many houses, however,

have experienced structural and nonstructural

damage that prevented immediate occupancy,

and resulted in expensive repairs.

Post-earthquake investigations have indicated

that much of the seismic damage can be attrib- uted to construction flaws or inadequate quality assurance, including missing fasteners, over- driving of nails, improper placement of anchor

bolts, and misplaced nails. The preference for more open interior spaces, hillside homes, soft stories due to garage openings, and more ir- regularly shaped structures may also be contributing factors. Poor soil conditions (liquefaction.

excessive settlement) or location (building on a

hillside) has been also cited as a cause of damage. Houses built in accordance with seismic

codes were generally found to perform well.

In general, poor seismic performance of woodframe houses has been observed when the house

did not respond as a unit, due to discontinuous

load paths. Common practices which resulted in the lack of integrity included:

• Insufficient or poorly detailed inter- component connections (anchorage to foundation, wall-to-wall connections, wall-toroof/floor connections).

• Inadequate bracing (cripple walls, total ab- sence of shear walls, large wall openings,

inadequate let-in bracing).

• Component separation (masonry fireplaces

and chimneys, masonry veneers, porch roofs

and other overhangs, different house sections).

• Non-uniform or irregular distribution of

stiffness, due to irregular plans and elevations (split level house, setbacks), which

leads to torsional movements.

• Poor detailing or quality of construction.

• Lack of continuous load path from roof to foundation.

In past hurricanes, the main cause of severe damage to wood-frame houses has been roof

damage, due to inadequate anchorage of the roof

framing to the wall, damage to the roof covering, and loss of roof sheathing. Foundation fail- ure and inadequate anchorage of the structure to the foundation have also caused failures. Similar to earthquake performance, issues such as continuous load path and structural integrity are

crucial for hurricane resistance, particularly be- cause water damage from roof failures is a major

contributor toward the economic losses. In addition to water damage, the internal pressure in a house is increased due to breaches in the building envelope from missiles, thereby contributing

to roof failures. Properly installed plywood

boards over windows or storm shutters, doors,

and garage doors can alleviate this problem, but

improperly installed boards and shutters can

334
contribute to the damage from wind-borne debris.

3 SUMMARY OF PAST EXPERIMENTAL

AND ANALYTICAL STUDIES

NIST has published a comprehensive state-of- the-art report on the structural performance of

single-family, wood-frame housing (Yancey et

al., 1998), which includes a detailed literature review of experimental and analytical studies

conducted on complete houses, in addition to wood subassemblies, such as shear wall diaphragms, and various connections. The following presents a summary of such experimental

and analytical studies. Additional details may

be found in the NIST report (Yancey et al., 1998).

3.1 Experimental Studies

Shear wall behavior under lateral loads has been

the focus of a significant amount of studies. Tests have been conducted under monotonic,

cyclic, and dynamic loads to investigate the re- sponse characteristics of the walls. Several additional studies have tested the sheathing-toframing connections to determine their response

characteristics, determine their resistance, and

provide data for analytical modeling; since they

have been shown to be a critical element in the

wall performance.

A relatively small number of full-scale houses

have been tested. Tests have been performed on

single- and two-story houses under field and

laboratory conditions. In addition, very few ex- perimental studies have been conducted on inter- component connections. Testing of those connections is important, since they are key ele- ments for the wood-frame building to act in an

integrated fashion and to provide data for computer modeling.

3.2 Analytical Studies

Analyzing the structural components of a woodframe building subjected to lateral loads is essential to understand their performance during

natural disasters. Since shear walls are the most

important elements in resisting lateral loads,

they have been extensively studied by several

investigators since the early 1970s. Simplified

analysis methods and finite element models

were developed to predict the nonlinear behavior

of the walls under static lateral loading. A few

researchers also attempted to predict the dynamic behavior of the walls subjected to earth- quake ground motion. Analytical models have

also been developed for floors and roofs under

uniform pressure.

The development of analytical tools for analyzing complete wood-frame houses started in the

mid 1980s. This work included analyzing the

different structural components, as well as the

inter-component connections in an assembled

model. Comparisons with experimental studies were performed to assess the validity of the pro- posed models and analyses. To date, only limited research has been directed toward the analysis of a complete building or the inter- component connections subjected to different

loading combinations.

4 OVERVIEW OF NIST RESEARCH

NIST has developed a research plan to improve

the state-of-the-art of wood-frame building

analysis tools, based on the current state of re- search and knowledge presented in the previous

two sections. Figure 1 shows a block diagram of

the technical approach of the research.

The objective of the NIST project is to develop

the metrics and predictive tools necessary for evaluating the structural performance of housing

systems during natural disasters, including

earthquakes and strong winds. Expected out- comes of this program include enhanced disaster

resistance and construction cost reduction by the

housing industry.

Using validated three-dimensional finite element

models of typical systems, structural perform- ance criteria for complete houses will be developed. The analytical model will include the in- teraction of the various diaphragms such as shear walls, floors, and roof, as well as inter- component connections. A library of the models

for the different diaphragms and connections

will be built. Data from prior testing conducted

335
on wood diaphragms will be used in the model.

As mentioned in the previous section, there is minimal data on the performance of inter- component connections, such as those between

wall and roof, wall and foundations, and inter- secting walls, therefore, this program will test the various connections under monotonic and

cyclic loads to characterize their performance

and provide the necessary data for modeling.

The developed models will be used not only for conventional wood-frame houses, but also for innovative materials and systems such as structural insulated panels (SIPs), insulated concrete forms (ICFs), and other systems that may be

suggested in the future. Testing will be con- ducted on new systems when modeling data are missing, and the models of their components and

connections will be added to the element library.

Once the three-dimensional model is assembled,

sensitivity studies will be conducted to establish

the baseline performance of conventional houses

and the performance of innovative systems. The

studies will use deterministic and stochastic pro- cedures to investigate the influence of various

parameters on the building performance. Those

parameters include building configuration, connection type, loading type, uncertainties in loading, and variations in material and connection properties. The sensitivity studies should

help establish performance criteria, determine

failure mechanisms, determine probability of

failure under a given loading, and identify weak

links and over-designed structural components.

Comparisons between the performance of conventional and innovative systems will be con- ducted to assess the merits of each system.

5 CURRENT EXPERIMENTS ON FULLSCALE WOOD-FRAMES

NIST is conducting experiments to obtain missing data on the performance of inter-component

connections. The first phase of experiments will

test wall-to-roof component connections similar

to those shown in Figure 2. The objectives of

this testing are to: 1) provide data for computer

modeling, 2) determine the level of uncertainty

in their characteristics, and 3) understand their

performance, since they have been identified as

a weak link in many structural failures. This

phase will include various types of wall-to-roof

connections such as toe-nailed; and hurricane

and seismic, ties and clips.

Prior to the component connection tests, eight

full-scale specimens similar to those in Figure 3 are currently being tested. The objectives of this testing are to:

1) understand the in-situ behavior of the wallto-roof connections in the structure,

2) determine the displacements and forces that will be applied to component connection

tests, 3) validate the three-dimensional models of

houses, and

4) explore possible failure mechanisms.

The specimens (see Figure 3) represent a center

cut through a simple wood-frame house. Each

specimen consists of two shear walls of 1 .22 m

(4 ft) length and 2.44 m (8 ft) height, supporting

four roof trusses spanning 4.88 m (16 ft). The

shear walls and roof trusses were purchased pre- fabricated from a local supplier. The shear walls

are supported on steel beams rigidly attached to the strong floor. The framing for each shear

wall consists of double top plates, single bottom

plate, and vertical studs spaced 0.40 m (16 in) on

center. All framing members are constructed

from 38 x 89 mm (2x4 nominal) spruce-pine-fir,

grade No, 2 lumber. 1 1 mm (7/16 in) thick ori- ented strand board (OSB) panels are used as outer wall sheathing and are connected to the

framing using 38 mm (L5 in) cement coated

staples spaced 76 mm (3 in) on perimeter and

1 52 mm (6 in) on intermediate studs.

Roof trusses, with a top chord slope of 4:12, are placed 0.40 m (16 in) apart. All truss elements

are 38 x 89 mm (2x4 nominal) members. 1

1

mm (7/16 in) thick OSB sheathing is attached to the trusses using 8d-common nails spaced 0.1 m

(4 in). Only toe nailed connections and hurri- cane clips are being used in these tests to tie the

roof trusses to the shear walls.

The specimens will be subjected to monotonic

uplift, monotonic lateral, and cyclic lateral loads. Uplift loads are applied using a vertical actuator with a steel loading tree that distributes the load

336
to ten points on the roof sheathing. Uplift point

loads on the roof sheathing are further distrib- uted using 2x8 lumber members. Lateral loads

will be applied using a horizontal actuator through steel pipes that are connected to the roof

trusses at their center of mass. Both the vertical and horizontal actuators push against a steel re- action frame. The specimens are highly instru- mented with displacement transducers and load

cells to gain maximum information on the rei sponse characteristics of the specimens,

j

'l At the time of this writing, only two specimens

have been tested under uplift load simulating

wind effects on structures. The first specimen

used a toe-nailed connection (two l6d-common

nails) to tie the bottom chord of the roof trusses

to the walls, while the second used a stronger

hurricane clip to connect the bottom chord of the

trusses to the upper plate of the walls' top plate.

For the toe-nailed connection, the specimen

failed at one side (south end) on the connection

between the trusses and the top plate of the south

wall. The failure mode is characterized as nail withdrawal from the top plate which resulted in separation in the bottom wood fibers of the bottom chord as can be seen in Figure 4. Figure 5 presents the uplift load-deformation curve for two of the failed connections.

For the specimen with the stronger hurricane

clip connection, failure was observed on one

side (north end) between the lower plate of the

walls' top plate and the wall studs (see Figure 6)

and also in the staples connecting the wall

sheathing to the same plate. This is due to the

high resistance of the hurricane clip between the

truss and the upper plate of the walls' top plate.

Figure 7 presents the uplift load-deformation curve for one of the failed connections.

Figures 5 and 7 indicate that the behavior of

both connections is highly nonlinear and that once the connections reach their ultimate load

they lose a significant portion of their resistance.

Comparing Figures 5 and 7, it can be observed

that the hurricane clip connection has a higher

uplift capacity and a larger failure deformation

than the toe-nailed connection. It can also be

seen that the residual strength of the hurricane

clip connection is equivalent to the capacity of

the toe-nailed connection.

6 CONCLUSIONS

NIST researchers are investigating the structural performance of housing systems. The perform- ance of wood-frame houses in past earthquakes

and hurricanes indicate that while recent disasters in the United States have resulted in few

casualties, they have caused substantial damage

and significant economic loss to residential

housing. An investigation of the state-of-the-art

in experimental and analytical research on wood-frame found that while significant progress has been made in analytical modeling and

testing of shear walls and diaphragms, only limited progress has been made on the testing and

modeling of complete houses and inter- component connections.

To help improve the performance of wood-frame

housing, NIST has started a research program to

fill some of the gaps in understanding their structural behavior during natural disasters. The

NIST research will include developing experimentally validated, three-dimensional, analytical

models to predict the performance of housing

units during natural disasters. This model will

be used to establish performance characteristics

of conventional and innovative housing systems.

Testing is being conducted on components for which modeling and performance data are lacking, such as inter-component connections. The

current experimental work on full-scale woodframes has shown that the behavior of wail-toroof connections is highly nonlinear and that the

strength of hurricane clips is much higher than

that of toe-nailed connections.

7 REFERENCES

Yancey, C. W., Cheok, G. S., Sadek, F., and

Mohraz, B. (1998), "A Summary of the Structural Performance of Single-Family, WoodFrame Housing," Technical Report NISTIR

6224, National Institute of Standards and Technology, Gaithersburg, MD.

337
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339
Figure 4. Failure of Toe-nailed Wall-to-Roof Connection

Force - Displacement Responses

Trusses 2 and 4, South End - Specimen #2 Test #3

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Figure 5. Load Deformation Curve of Toe-nailed Wall-to-Roof Connections

340
Figure 6. Failure of Wall-to-Roof Connection with a hurricane clip

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341

Performance-Based Damage Assessment and Design Verification

of U.S. Housing in Extreme Earthquake and Hurricane Events

Jay H. Crandeir and Will iam Freebome'

ABSTRACT

This paper presents an evaluation of U.S.

residential construction with respect to the actual

performance of single-family detached

dwellings subjected to extreme earthquakes and

hurricanes. Design and construction implications

are also addressed in a scientifically rigorous

approach based on statistical sampling of the

affected housing populations, engineering

analysis, and fragility modeling. Such an approach is believed to be an effective means to identify performance issues, evaluate the

strength of cause-and-effect relationships, and

properly focus attention on cost-effective

solutions to performance problems that will

improve future design methods and construction

practices.

KEYWORDS: Houses; hurricanes; earthquake;

performance; design; assessment; damage;

statistics; fragility.

1.0 INTRODUCTION

In this study, the "experience" of conventional

residential construction (light-frame wood) is evaluated with respect to single-family detached

(SFD) housing performance in the Northridge

Earthquake and Hurricane Andrew using a

scientifically sound experimental approach.

Major seismic events cause damage to lateral resisting wall systems in conventional wood-

'Director, Structures and Materials Division, NAHB

Research Center, Inc., 400 Prince George's Blvd,

Upper Marlboro, MD 20774-8731 (www.nahbrc.org)

'Mechanical Engineer, Office of Policy Development

and Research, U.S. Department of Housing and

Urban Development, 451 Seventh Street, S.W. - Room 8132, Washington, DC 20410 (www.hud.gov)

frame construction and major wind events often cause damage to roof systems. Since these

systems tend to predominate the structural risk

to homes in earthquakes and hurricanes,

understanding their actual performance is essential to understanding the overall

performance of residential construction.

The objective of this paper is to evaluate the

actual performance of homes during the

Northridge Earthquake and Hurricane Andrew

using statistical data on key factors such as construction characteristics, damage frequency,

ground motions, wind speed, and system

strength estimation. The current approaches to seismic design and wind design are also

evaluated by investigating the ability to predict

the actual damage and explain certain

fundamental cause-and-effect relationships.

2.0 EVENT CHARACTERIZATIONS

2.1 Northridge Earthquake

The Northridge Earthquake occiured at 4:30

a.m. on January 17, 1994. Its epicenter was

located in a densely populated area of Los

Angeles County near the community of

Northridge. Over 30 deaths were reported as a

direct result of the tremor, and a total death toll of 58 was attributed to both direct and indirect causes [1]. Fortunately, relatively few deaths were associated with single family dwelling

construction even though most people were in their homes at the time of the tremor. Overall

property loss estimates range from $20 to $30

billion. While this earthquake was the most

costly natural disaster experienced in the United

States, its magnitude (Richter Scale Ml = 6.4,

Surface Wave Magnitude Ms = 6.8, Moment

Magnitude Mw = 6.7) was modest in terms of

other more severe earthquakes. The Northridge

Earthquake produced an unprecedented set of

by

343
strong motion records, including more than 250

records considered representative of free field ground motion [2]. The recorded strong ground

motions were characterized by a large degree of

variation between sites at comparable distances

from the source. Several different causes of

spatial variation in ground motions have been

recognized which are systematic and predictable

to some extent. Strong ground motions are influenced by the characteristics of the seismic

radiation from the source, by the process of wave propagation between the earthquake

source and the recording site, and by local site effects [3],

2.2 Humcane Andrew

Hurricane Andrew struck a densely populated

area of southern Florida on August 24, 1992.

with peak winds in excess of 78.2 m/s (175

mph) recorded at one location [4]. Using

advanced hurricane modeling techniques and

available wind speed data, the estimated over- land wind field of Hurricane Andrew was

modeled as shown in Figure 1 [5]. The winds

produced by Hurricane Andrew were estimated

to have mean recurrence interval of about 300

years [6]. Because of the distance of the

housing stock from the shoreline, most of the

damage was related to wind, rain, and windbome debris —not storm surge.

3.0 DAMAGE ASSESSMENT

METHODOLOGY

The objectivity of the damage assessment method

hinges on the use of sampling procedures to gain a valid representation of the performance of the

population of affected homes and to avoid bias

toward less severe or more severe occurrences of

damage. This report focuses on the study of

single-family detached (SFD) housing. The

sampling method and damage rating approach are

briefly described below. For a more detailed

treatment, the reader is referred to the original

studies [7][8].

For the Northridge Earthquake, a 16. 1 km (10 mi)

radius around the epicenter was selected as a damage zone for study. Postal regions—as defmed by five-digit zip codes that fell within the

damage zone or intersected its border—were used

as the basis for a random selection of homes from

property tax records. There were 183,514 SFD

home entries in the property tax database for the designated zip codes. Seventy-five sites were randomly selected by street address from the tax record database. The home at the selected address was surveyed along with two homes on either

side—a total of five homes p>er site. The survey

teams visited 75 sites resulting in 341 usable

survey forms.

For Hurricane Andrew, over 600 randomly

sampled homes experiencing the highest winds were assessed using detailed field survey forms.

The randomly determined damage study

locations are shown in Figure 1. The sample was drawn from a street map. In the first stage, a random sample of map grid areas was made.

Within the sampled grids, a random sample of

streets was drawn. All homes on the sampled

streets were surveyed, with exception of those

homes on a street that were not surveyed at the

end of a day. A total of 466 survey forms were deemed suitable for analysis.

In each survey, all of the sample homes were

visited by one of several assessment teams to document the housing characteristics and the

level of performance. Survey forms were used

to document about 60 construction

characteristics and 30 damage characteristics for

each home. Photographs were taken of each

assessed property. For each category listed on the

siirvey fonn, earthquake or wind damage was

graded according to four levels of severity:

• NONE: no visible damage;

• LOW: components are stressed, but in functional condition (i.e., minor cracking of

stucco or minor loss of roof sheathing); • MODERATE: evidence of severe stress, permanent deflection, or near failure in any

structural system (i.e., severe cracking of

stucco or loss of many roof sheathing panels

without collapse of the structure or roof

system); and,

• HIGH: partial or complete failure (i.e., collapse) of any structural system

344
Case studies of rare instances of extreme damage

were also conducted for each event, but they are beyond the scope of this paper. Similarly,

statistical surveys and case studies were also

performed for multi-family low-rise and singlefamily attached forms of residential construction.

4.0 CONSTRUCTION CHARACTERISTICS

AND DAMAGE STATISTICS

4. 1 Northridge Earthquake Survey

The SFD housing characteristics are briefly

summarized in Table 1. Using information

provided by the property tax record database and

inspectors with the Department of Building and

Safety, it was determined that about 90 percent of

the homes in the sample were built prior to the 1971 San Fernando Valley Earthquake when

simple prescriptive requirements were normal to SFD home construction. About 60 percent of the surveyed homes were built during the 1950s and

1960s. House age ranged from the 1920s to the

early 1990s. SFD homes were typically one-stor}'

and nearly two-thirds had an attached garage.

Styles of SFD homes ranged from expensive

custom homes to affordable, older homes. As

expected, all homes surveyed had wood exterior

wall framing and most did not use structural sheathing for wall bracing. Instead, wood let-in braces, Portland cement stucco, and interior wall

finishes provided lateral resistance. Homes on

crawlspace foundations outnumbered those on

concrete slabs by almost rwo-to-one, despite a

notable increase in the use of slab-on-grade

foundations since the 1960s, Most of the

crawlspace foundations used full-height concrete

or masonry stem walls, not cripple walls.

The performance of SFD homes is shown in Table 2. The table is broken into observed

damage of the homes sampled and estimates of

damage for the entire population of homes within

the survey area. Confidence intervals at the 95

percent level are shown for each estimate.

Damage to structural elements—foundation, wall

firaming, and roof firaming—was limited to a small

proportion of surveyed homes. In general, SFD

homes suffered minimal damage to elements that are critical to the safety of occupants. Of the

structural elements, damage was most common in the foundation system. The small percentage of

surveyed homes that experienced moderate to high foundation damage were located in areas that endured localized ground effects or problems

associated with hillside sites. The localized

ground effects included fissures or ground

settlement that cracked foundations. For hillside

sites, partial slope failures contributed to the foundation damage.

Interior and exterior finishes suffered more

widespread damage than foundation and framing,

with only about half the buildings escaping

unscathed. However, the great majority of

damage was limited to the lowest rating

categories. Stucco was observed on nearly all home exteriors. Damage to stucco usually

appeared as hairline cracks radiating from the comers of openings—particularly larger openings

such as garage doors—or along the top of the foundation. Interior finish damage paralleled the occurrence of exterior finish (stucco) damage.

Resilient finishes—such as wood panel or lap

board siding—fared well and often showed no

evidence of damage even when stucco on other

areas of the same building was moderately

damaged.

The Chi-square test was used—at a 95 percent

level of confidence—to judge statistical significance of various conditions on the outcome

of a home's performance. The inferences initially

designated for study by the Chi-square test included observed performance versus:

• peak ground acceleration estimates,

• age of the home,

• roof type

• number of stories, and

• foundation type.

Chi-square analyses requires a large number of

observations in each category to produce valid

results. Thus, the analysis was limited to the exterior damage rating as the performance

indicator since it represented the greatest extent of

damage. Also, the low, moderate, and high

damage ratings were grouped such that a

345
damage/no-damage test was applied. Except for foundation type, all inferences were inconclusive.

Using only data from one-story homes,

comparison of crawlspace versus slab-on-grade

foundation construction shows a significant

difference in the level of damage to the stucco used on one-story homes. Single-story homes

with slab foundations exhibited damage to exterior finishes in about 30 percent of the cases, while homes on crawlspace foundations with

masonry or concrete stem walls approached a 60

percent rate of occurrence. Since the majority of

crawlspace homes in the survey area were built before 1960 while nearly three-quarters of slab

homes were built after 1960, stucco performance

alone may not be sufficient to conclude that one

foundation type is necessarily better than another.

There may be factors influencing stucco performance other than merely foundation type,

such as house age.

Case studies of damage were conducted on 54

SFD homes that experienced rare, severe damage.

The most notable sources of structural damage to these case study homes were related to ground

conditions (e.g., fissures and settlement) and

hillside construction conditions (e.g., weak

foundation connections or partial slope failures).

Damage to wall finishes, contents, mechanical

equipment, masonry chimneys, and masonry

privacy fences was much more common.

4.2 Hurricane Andrew Survey

Table 3 summarizes the key construction

characteristics of the sampled homes in the

Hurricane Andrew study. Most of the homes

were one story in height with nominally

reinforced masonry walls, wood-framed gable

roofs, and composition shingle roofing.

Table 4 summarizes the key damage statistics determined for the sampled homes in Hurricane

Andrew. As expected, the most frequent form

of damage was related to windows and roofmg

with 77 percent of the sampled homes suffering

significant damage to roof covering materials.

Window damage resulting in at least one broken

window was realized in 91 percent of the

sampled homes. Blown-off roof sheathing was

the most significant aspect of the structural damage with 64 percent of the homes losing one

or more roof sheathing panels. Loss of windows

and roofing in Hurricane Andrew led to widespread and costly water damage to interiors.

In a similar study of Hunricane Opal, wind

speeds ranged from about 44.7 m/s to 51.4 m/s

(100 to 115 mph) based on peak gusts at a 10m

(33 ft.) height and normalized to open terrain over the sample region. Again, roofing damage

was the most common form of wind damage, but

at a frequency of only 4 percent of the housing

stock [9]. Roof sheathing damage was realized

in less than 2 percent of the affected housing

stock. This data provides a good contrast to that obtained from the Hurricane Andrew study.

Aside from the much lower wind speeds in Hurricane Opal, most of the homes were

shielded by trees whereas those in South Florida were in typical suburban residential exposure

(wind exposure B) and trees, when present, were denuded in the extreme winds of Andrew. The

more extensive damage in Hurricane Opal was

caused by storm surge to homes and other

buildings on the barrier islands.

Some inferences on the Hurricane Andrew data were found to be statistically significant at a 95

percent confidence level. For example, gable

roofed homes suffered significantly higher

damage on average than their hip-roofed

counterparts with respect to the amount of roof

damage realized. It should be noted that this difference in vulnerability is not reflected by

commensurate differences in wind loads

calculated for roof components in ASCE 7 [10].

Also, two-story homes experienced significantly

higher average damage than one-story homes

with respect to window and water damage, but

not roof damage.

The level of damage to walls was very low with

about 2 percent of the homes experiencing some

significant form of wall structural damage.

About 96 percent of the homes were constructed

of nominally reinforced masonry walls (i.e. #4

vertical rebar at 2.4 m (8 feet) on center).

Interestingly, this amount of reinforcement falls below the minimum reinforcement ratios that are required in current reinforced masonry or

346
concrete design specifications in the United

States. Similarly, roof tie-down failures were

reasonably low with 84 percent of the sampled

homes experiencing no form of failure to these

important elements in high wind regions. About

8 percent of the homes experienced a partial or

full roof blow-off failure while most homes had

breached envelopes (i.e., broken window glazing

and/or a failed garage door) resulting in a higher

internal pressure condition according to current design practice represented in ASCE 7.

5.0 STRUCTURAL PERFORMANCE

EVALUATION

5.1 Northridge Earthquake

The findings reported herein are summarized

from a more thorough evaluation and report on

the Northridge Earthquake damage statistics [11]. A total of 49 homes were extracted from

the Northridge Earthquake survey discussed

previously which met the following criteria:

• one-story, single-family homes;

• stucco exterior wall finish (without

structural panel sheathing);

• asphalt composition roof shingles; and

• adequate photographic documentation to characterize the street facing wall

configuration.

These criteria resulted in a fairly homogenous,

random sample for the purpose of evaluating

fundamental relationships between damage and

various seismic design factors. The application

of findings from this study is limited to the types

of homes that meet the above criteria. These

homes were of typical construction in the San

Fernando Valley with stucco and wood let-in wall bracing. Roughly 80 percent of these

homes had crawlspace foundations and 20

percent had slab-on-grade foundations.

Table 5 summarizes the damage to the exterior

wall finish of the 49 homes extracted from the survey. The percentages in Table 5 closely

coincide to the damage statistics for the

complete survey sample of the 341 homes as shown in Table 2.

5.1.1 Solid Wall Ratio (P) vs. Stucco Damage

Rating

The value of the solid wall ratio, p, was

determined from the photographic

documentation of the street-facing side of the

sampled houses. It is a simple ratio of the length

of solid wall segments (i.e., without openings for windows and doors) to that of the entire wall

line. Figure 2 indicates that there is no obvious

correlation between the amount of solid stucco

wall and the damage rating for the homes in this study. This finding was unexpected. However,

there is a possible slight trend when looking at the average p for the NONE and LOW

groupings which are 52 percent and 49 percent,

respectively. Unfortunately, the variability

(scatter) of the data obscures any statistically

conclusive finding.

5.1.2 Wall Discontinuities vs. Stucco Damage

Rating

Both NEHRP - 97 and UBC - 97 include design

provisions for homes constructed with plan

irregularities [12] [13]. Many of the homes in

this study had an Out-of-Plane Offset plan

structural irregularity. By definition, an out-ofplane offset is a discontinuity in a lateral force

resistance path, such as out-of-plane offsets of

the vertical elements. This type of irregularity is believed to be associated with substandard

performance and current residential construction

codes limit a braced wall line offset to 1.2 m (4

ft) for this reason. However, Figure 3 shows no

apparent correlation between the number of wall

discontinuities on the street facing wall and

damage rating for the homes in this study.

Figure 4 considers only offsets equal to or

greater than 1.2 m (4 ft) as a wall discontinuity.

Again, there seems to be no apparent correlation

between the stucco damage rating and the

number of offsets greater than or equal to L2 m

(4 ft).

5.1.3 Spectral Response Acceleration vs. Stucco

Damage Rating

Data from 9 strong motion stations that recorded

ground motions during the Northridge

347
Earthquake were examined in this study. These

9 stations were in close proximity to the 49

homes investigated in this study. The peak

horizontal component for the 0.2 second

acceleration response spectra and the 1.0 second

acceleration response spectra for 5 percent

critical damping were extracted from the data

sets for the 9 strong motion ground stations. This data is summarized in Table 6 along with

the corresponding return period estimate. The

return periods were determined using the U.S.

Geological Survey (USGS) hazard curves for each site. The methodology used to create seismic maps and hazard curves is discussed in other reports [14][15].

Ground motion amplification has been observed

at the Tarzana - Cedar Hill station in many

earthquakes for both strong and weak ground

motions. The Tarzana station is located near the

crest of a low (20 m) natural hill on the south

side of the San Fernando Valley. Topographic

effects can explain many features of the

observed amplification patterns, but the three

dimensional geological structure beneath the hill may also be responsible in part for the very large

observed amplification at the top of the hill [16].

For this reason the ground motions at Tarzana were considered anomalous and not used in this study.

The latitude and longitude for each of the 49

houses extracted from the survey were

determined. The distance of the homes to the nearest strong motion stations was determined

from the latitude and longitude coordinates. The

corresponding acceleration response spectra was

determined by interpolating between the ground

motion stations closest to the house of interest. Figure 5 graphically indicates that there was no

apparent correlation between the 0.2 second

spectral response acceleration and the stucco damage rating for the sampled homes. Figure 6

suggests that there may be a slight trend between

the 1 .0 second spectral response acceleration and

damage rating on average. The average 1.0 second spectral response acceleration for the

NONE and LOW groupings are 40 percent and

46 percent, respectively. It is impractical to consider if the trend is statistically significant

since the sample is relatively small and the

scatter is relatively large. Although a weakly

supported observation. Figure 6 provides an

indication that the longer period spectral

response acceleration may be the better seismic

design parameter for small conventionally built wood-framed buildings such as homes since it appears to better explain the damage.

5.1.4 Case Study - Mecca Avenue

Table 7 summarizes three homes surveyed on Mecca Avenue that were part of the random

sample of homes. They were of very similar

construction (style) and oriented identically to the ground motion. The three homes

summarized in this case study reinforce some of

the findings discussed earlier in this paper. The

surveyed homes were on the same street and of

close proximity to the Tarzana strong motion

station. The Tarzana station realized some of the

largest ground motion readings during the

Northridge Earthquake. In this anecdotal

comparison, the damage rating improved with

increasing wall discontinuity and decreasing

solid wall amount. This finding, though not

statistically conclusive, is exactly opposite of

conventional engineering theory.

5.2 Hurricane Andrew

The performance of roof sheathing components

in Hurricane Andrew was evaluated using

principles of engineering and risk modeling and

this report summarizes the major findings from

the original study [17]. To facilitate the

perfonmance evaluation, a Monte Carlo

simulation model was developed to predict the

frequency of roof sheathing damage (i.e., estimate the percentage of homes experiencing

the loss of at least one panel of roof sheathing).

The model accounted for variation in sheathing

fasteners, construction materials (i.e., framing

lumber species and density). South Florida

housing characteristics, workmanship effects,

and other important parameters affecting the

actual sheathing resistance values and wind

loads (i.e., surface pressures) experienced. The

sheathing pull-off resistance values and variation

used in the model were based on tested data with

adjustments to account for roof framing lumber

used in South Rorida [18]. The wind speed over

348
the sample region (Figure 1) was essentially

treated as a deterministic parameter and was

based on the modeled range of wind speeds

experienced over the sample region (i.e., mean =

72.6 m/s (162.5 mph), COV = 0.01). The wind

load provisions of ASCE 7 were used to determine wind pressures on the roof sheathing

using a suburban exposure, enclosed building

internal pressure condition (i.e., applicable to the

attic space prior to sheathing loss), and a wind

directionality factor of 0.85.

From the statistical damage survey data

discussed previously [8], a sub-set of singlestory homes with gable roofs and composition

shingle roofing was selected as a homogenous

sample for the performance evaluation and

model calibration. Of these randomly sampled

homes, 69 percent±6 percent (95 percent

confidence limits) experienced the loss of at least one panel of roof sheathing, and all were

in the 71.5-73.8 m/s (160-165 mph)(peak gust)

region of Hurricane Andrew as shown in Figure

1.

The key results of the study are the fragility curves shown in Figure 7 which give predictions

of the percentage of homes expected to experience the loss of one or more panels of roof

sheathing based on the hurricane magnitude

(wind speed). Figure 7 contains four different

fragility curves representing different levels of

roof sheathing attachment. The curve

representing the actual housing population

(second from the top) is calibrated to the actual

observed roof sheathing damage frequency of 69

percent±6 percent (95 percent confidence limits)

at a wind speed of 72.4 m/s (162 mph) (peak

gust) with a mix of 6d and 8d sheathing nails

representing roof sheathing attachments in the

modeled housing population. The model

prediction is 71 percent — tending toward a

slight over-estimation of the actual damage

frequency.

Some very interesting findings are drawn from

the analysis:

• Actual Performance Better Than Minimum

Code

The minimum requirement for roof sheathing

attachment in the 1991 South Florida Building

Code was 6d cornmon nails spaced at 30 cm (12

in) on center in the field of the roof sheathing

panels (see top curve in Figure 7) [19]. If this minimum requirement was representative of the

entire housing population, the predicted

frequency of roof sheathing loss would have

been about 92 percent instead of 69 percent of

the homes. One may conclude that this minimum code requirement was very

insufficient for the South Florida wind climate

and that the housing performance was actually

better than that implied by the code minimum.

However, neither the code-implied performance

nor the actual performance is considered

acceptable.

• Revised Code Minimum Will Significantly

Improve Performance

The next most important observation is that the

level of performance expected for new homes

constructed under the 1994 South Florida

Building Code will be significantly improved

(bottom curve in Figure 7) [20]. This

improvement is primarily attributed to increasing the roof sheathing attachment

requirement to 8d common nails from 6d

common nails without any change in workmanship or inspection that may result in improved installation quality. Decreasing the

spacing to 15 cm (6 in) on center and requiring

sheathing attached to the gable end framing to be nailed at 10 cm (4 in) on center are also

important features in the new code. New homes

constructed under this revised code provision are predicted to perform with a roof sheathing loss frequency of about 1 percent in the next event

equivalent to Hurricane Andrew.

• Benchmark for Acceptable Performance

An acceptable roof sheathing damage frequency

may be targeted at about 10 percent of the

homes for an event the magnitude of Hurricane

Andrew in the South Rorida wind climate (see

lower box in Figure 7 at 72.4 m/s (162 mph)

wind speed). This target is risk-consistent with

the generally acceptable performance of

standard roof sheathing attachments in typical

349
extra-tropical wind climates covering most of

the non-coastal United States. Thus, a damage

frequency of about 10 percent of the homes

losing one or more panels of roof sheathing in an

event similar to Hurricane Andrew constitutes a respectable goal or target performance for the

South Florida housing population. Obviously,

the actual frequency of roof sheathing damage

(i.e., 64 percent of the all SFD homes) does not meet the proposed benchmark for acceptable

performance.

Though not similarly evaluated, the incidence of

building collapse (wall racking) or roof blow-off

should have a lower target damage frequency

because of the more severe consequences.

Indeed, the occurrence of this type of damage

was documented at a reasonably low level in Hurricane Andrew - 2 percent for wall damage

and 8 percent for roof-wall connection damage

[8]. (The 2 percent wall damage statistic is primarily associated with wood frame wall

construction which comprised about 4 percent of

the sampled homes). Of the nominally

reinforced masonry homes sampled, no wall

failures were documented; extremely rare incidences of severe masonry wall damage,

however, were found in the housing population

[8].

• Possible Wind Directionality and Shielding

Effects

One point of concern following the generation of

the fragility curves was related to the level of

damage at lower wind conditions. According to the wind map in Figure 1, the maximum wind

speeds for Hurricane Andrew were less than

about 62.6 m/s (140 mph) for areas north of 88''^

Street (Kendall Ave.). Kendall Avenue was the

northern boundary for the statistical sampling

region in the HUD damage survey [8]. This

northern boundary of the 'damage zone' was

selected because of the lack of any significant

damage, particularly roof sheathing loss,

proceeding further northward.

As shown in Figure 7, the calibrated MCS

model predicts damage frequencies of more than

50 percent at the 62.6 m/s (140 mph) wind

speed. One possible reason is that the rate of

wind speed decrease in Figure 1 is too low north

of the eyewall of Hurricane Andrew (i.e., an

error in the hurricane. wind field model). This

would require that the wind speeds drop from

71.5 m/s (160 mph) to less than about 44.7 m/s

(100 mph) at the crossing of Kendall Avenue to agree with a damage level well below 10 percent

based on observation in this region (see Figure

7). This wind-speed drop-off may be physically

unreasonable, though not implausible.

Therefore, it is believed that this drop-off in actual damage relative to the modeled damage is at least in part due to a change in the wind

exposure condition (exposure B was used in the model). This change in wind exposure is attributable to two characteristics. First, the areas north of Kendall Avenue were generally

older developments with more mature trees which could have shielded the lower buildings

(i.e., one-story homes). Second, the wind speeds

in this region were low enough that the trees were able to maintain some level of protection

during the hurricane without being completely

destroyed or denuded. It is also possible that the

generally older homes in this region were built using materials and methods that were somewhat more wind resistant.

To test this hypothesis, the MCS model was re- run at the 62.6 m/s (140 mph) wind speed with an additional adjustment to account for possible

shielding effects due to the survival of the trees. The calculated wind loads were reduced by 25

percent for shielding which corresponds with the

results of a recent wind tunnel study of low-rise

structures in a densely built-up (i.e., shielded),

exposure B environment [21]. With this adjustment, the MCS model predicted a roof

sheathing damage frequency of about 1 percent

which agrees reasonably well with the anecdotal

observation of very little structural damage north

of Kendall Avenue. Thus, the existence of

mature trees in a relatively dense mix with

development may reduce wind damage

significantly, provided the wind speeds do not exceed the ability of the trees to act as barriers.

Some of this disparity may be attributed to an unaccounted increase in wind directionality

effect not considered in the model. Proceeding

north of the hurricane eyewall, the wind speed

350
magnitude and range of wind directions

experienced are both diminished. Thus, a wind

directionaUty factor of 0.75 (as applicable to extra-tropical wind conditions) may be

appropriate as concluded in an extensive

analytical wind tunnel study [21].

It is probable that the following factors all contributed to an apparent over-prediction of

damage north of the eyewall (outside of the

statistical damage study region):

• wind field modeling errors,

• increased reductions in wind load

associated with wind directionality, • increased shielding associated with the

survival of trees; and,

• differences in the construction materials

and methods used in the generally older

homes located north of the eyewall and

the sample study region.

6.0 CONCLUSIONS

6. 1 General

1. It is feasible and economical to obtain

reliable and representative data on housing

construction and damage frequencies (i.e., performance) following extreme natural events such as earthquakes and hurricanes.

2. Construction and performance data collected

and evaluated as described in this paper may

be used to rationally assess performance,

identify need for improvement, evaluate the

adequacy of engineering methods relative to

actual performance, and develop costeffective solutions related to the real problems associated with natural hazards.

6.2 Northridge Earthquake

1. The single-family housing population

performed reasonably well in the Northridge

Earthquake with 95.9 percent of the sampled

homes receiving a stucco damage rating of

NONE or LOW; very few homes exhibited

significant structural damage related to life- safety concerns.

2. Despite fundamental structural performance

relationships assumed in conventional

seismic design practices and theory, there was no obvious or significant correlation

between solid wall ratio or spectral response

acceleration and the damage rating for the

statistically representative sample of homes

examined in this study.

3. The findings provide some indication that the long period ( 1 second) spectral response

acceleration is the preferable ground motion

parameter for design of small, light-frame

structures.

4. This study also confirms that certain

irregularities, such as wall out-of-plane

offsets, do not have the affect on performance implied by current design and

construction provisions.

6.3 Hurricane Andrew

1. Given the magnitude of Hurricane Andrew,

the structural (life-safety) performance of

the typical South Florida housing stock (i.e., masonry walls with wood frame roofs) was

found to be very reasonable with the

prominent exception of roof sheathing

attachment.

2. Wind loads determined in accordance with

ASCE 7 provide a reasonable basis to assess the actual performance of residential roof

components given the use of an "enclosed

building" internal pressure condition for the

determination of roof sheathing loads, the use of a 0.85 wind directionality factor, and

appropriate consideration of wind shielding

effects.

3. The actual performance of the housing stock was, on average, better than that implied by

the governing building code provisions

preceding Hurricane Andrew's devastation -

though both are considered unacceptable.

4. Revisions to the South Rorida Building

Code (SFBC) following Hurricane Andrew

resulted in roof sheathing attachment

provisions that are slightly conservative, but

practical and effective.

351
5. The statistical data on roof-to-wall

connection performance in Hurricane

Andrew gives ample evidence that the roof

tie-down connection (as required by SFBC

and executed in practice) was reasonably

reliable.

7.0 REFERENCES

[1] National Institute of Standards and

Technology, 1994 Northridge

Earthquake - Performance of

Structures, Lifelines, and Fire

Protection Systems, Gaithersburg, MD,

1994

[2] Grazier, V.M., Strong Motion Data from

Ground-Response Sites and Structures

Recorded During the 1994 Northridge

Earthquake, Proceedings Northridge

Earthquake Research Conference, Los

Angeles, California, 1997.

[3] Somerville, P., Research Overview on

Strong Ground Motion: The 1994

Northridge Earthquake, Proceedings of

the NEHRP Conference and Workshop

on Research on the Northridge,

California Earthquake of January 17,

1994, Richmond, CA, 1998.

[4] Reinhold, T. A., Vickery, P. J., and

Powell, M. D. (1993). "Windspeeds in Hurricane Andrew: myths and reality."

7th U.S. Nat. Conf. on Wind Engrg.,

Univ. of California, Los Angeles, Calif.,

553-562.

[5] Vickery, P.J., P.F. Skerlj, A.C. Steckley,

and L.A. Twisdale, Hurricane Wind

Field and Gust Factor Models for Use

in Hurricane Wind Speed Simulations,

Applied Research Associates, Raleigh,

NC, 1997.

[6] Vickery, P.J., P.F. Skerlj, A.C. Steckley,

and L.A. Twisdale. Simulation of

Hurricane Risk in the United States Using an Empirical Storm Track

Modeling Technique, Applied Research

Associates, Raleigh, NC, 1998. (pending

journal publication).

[7] Assessment of Damage to Residential

Buildings Caused by the Northridge

Earthquake, Prepared for the U.S.

Department of Housing and Urban

Development by the NAHB Research

Center, Inc., Upper Marlboro, MD,

1994.

[8] Assessment of Damage to Single-Family

Homes Caused by Hurricanes Andrew

and Iniki, Prepared for the U.S.

Department of Housing and Urban

Development by the NAHB Research

Center, Inc., Upper Marlboro, MD,

1993.

[9] Assessment of Damage to Homes

Caused by Hurricane Opal, Prepared

for the Florida State Home Builder's

Association by the NAHB Research

Center, Inc., Upper Marlboro, MD,

1996.

[10] Minimum Design Loads for Building

and Other Structures (ASCE 7-95),

American Society of Civil Engineers,

New York, NY, 1995.

[11] Evaluation of Housing Performance and

Seismic Design Implications in the

Northridge Earthquake, Prepared for the

U.S. Department of Housing and Urban

Development by the NAHB Research

Center, Inc., Upper Marlboro, MD, June

1999.

[12] NEHRP Recommended Provisions for

Seismic Regulations for New Buildings

arui Other Structures, Building Seismic

Safety Council, Washington, D.C.,

1997.

[13] Uniform Building Code: Volume 2,

International Conference of Building

Officials, Whittier, CA, 1997.

[14] Frankel, A., Mapping Seismic Hazard in the Central and Eastern United States,

352
Seismological Research Letters, Vol. 66,

No. 4, pp. 8-21, 1995.

[15] Frankel, A., Mueller, C, Bamhard, T.,

Perkins, E., Leyendecker, E., Dickman,

N., Hanson, S., Hopper, M., National

Seismic Hazard Maps, June 1996:

Documentation, U.S. Geological

Survey, Open File Report 96-532, 1997.

[16] Darragh, R.B., Graizer, V.M., Shakal,

A.F., Tarzana, California: Site

Response and Characterization,

Proceedings of the NEHRP Conference

and Workshop on Research on the

Northridge, California Earthquake of

January 17. 1994, Richmond, CA, 1998.

[17] Reliability of Conventional Residential

Construction: An Assessment of Roof

Component Performance in Hurricane

Andrew and Typical Wind Regions of

the United States, Prepared for the U.S.

Department of Housing and Urban

Development by the NAHB Research

Center, Inc., Upper Marlboro, MD,

1999.

[18] Mizzell, D.P. and Schiff, S.D., Wind

Resistance of Sheathing for Residential

Roofs, Clemson University, Clemson,

SC, 1994.

[19] South Florida Building Code, Dade

County, FL, 1991.

[20] South Florida Building Code, Dade

County, FL, 1994.

[21] Ho, Tat Chiu Eric, Variability of Low

Building Wind Loads, Ph.D. Thesis,

Department of Civil Engineering, The

University of Western Ontario, July

1992.

353
Figure 1 Maximum 3-Second Gust Wind Speeds (mph)

Experienced in Hurricane Andrew at 10 m (33 ft) Elevation over Open Terrain

(Courtesy Applied Research Associates, Raleigh, NC)

1 mph = 0.447 m/s

0

0

0

.2 0

CO

=:

CO ^

1/3

.90

80

70

60

50

40

30

20

10

00

NONE

JiL.

4

4

1 LOW

Damage Rating

MODERATE

Figure 2

Solid Wall Ratio (P) vs. Damage Rating

(Data Points Represent Individual House Samples)

354
3

uo

18

16

14

12

10

86

4

20

No

N

1

i 1

1 Offset Offset Offsets Offsets

Damage Rating

NONE

SLOW

MODERATE

Figure 3

Non-Correlation of Wall Offsets and Damage Rating

Damage Rating

NONE

BLOW

MODERATE

1 2 3 Offset Offsets Offsets

Figure 4

Non-Correlation of Wall Offsets Greater than 1.2 m (4 ft) and Damage Rating

355
60%

c

c. CD c

a: o u

L.

CJ < o

50%

40%

r, '« 30%

20%

10%

0%

t t

NONE LOW MODERATE

Damage Rating

Figure 5

0.2 Sec Acceleration Response Spectra vs. Damage Rating

(Data Points Represent Individual House Samples)

100%

90%

80%

70%

60%

50%

40%

30%

20%

10%

0%

NONE LOW MODERATE

Damage Rating

Figure 6

1.0 Sec Acceleration Response Spectra vs. Damage Rating

(Data Points Represent Individual House Samples)

356
Figure 7

Roof Sheathing Fragility Curves

for One-Story Homes in a Suburban Exposure

(1 mph = 0.447 m/s).

357
Table 1 Construction Characteristics of SFD Dwellings

COMPONENT NORTHRIDGE HOMES, 1994

Ntimhpr of StnriP's 19% One

18% Two

1% Other

Wall Sheathino 8(1% None

7% Plvwood

1 T Inknnwn

Pni inri?i t ion Tvnp

34% Slab

8% Other

Exterior Finish 50% Stucco/Mix

45% Stucco Only

6% Other

Interior Finish 60% Plaster 26% Gypsum Board

14% OtherAJnknown

358
Table 2

Description of Damage to Single-Family Detached Homes

(Northridge Earthquake)

Observed

Damage

Sample

Size No Damage Low Damage Moderate Damage High Damage

Foundation 327 295 26 3 3

Foundation-to- Walls 324 293 24 5 2

Walls 317 311 6 0 0

Roof 328 326 2 0 0

Exlenor Finish 306 155 141 9 1

Intenor Finish 265 132 1 22 1

1

0

la ' cf 1 mo t Ai/H II o m<i o o jLollIUalcll ivdllUi^c within Survey Area No Damage Low Damage Moderate Damage High Damage

Foundation 87.9% < 90.2% < 92.59c 5.9% < 8.0% < 10.0% 0.3% < 0.9% < 2.7%. 0.3% < 0.9% < 2.7%

Foundation-to-WaJls 88 1%<90.4% <92.7'7c 5.4% < 7.4% < 9.4% 0.5% < 1.5% < 3.6% 0.2% <0.6%> <2.2%

Walls 94.0% < 98.1% <99 0% 0.9%< 1.9%p <4.1% 0.0% < 0.0% <0 9%- 0 0%<0.0%; <0.9%

Roof 97.2% < 99.4% < 99.6% 0.2% < 0.6% < 2.2% 0.0% < 0.0% < 0.9% 0.0% <0.0%. <0.9%P

Exterior Finish 46.7% < 50.7% <54.7%c 42.1% < 46.1% < 50.1% 1.6% < 2.9%. < 5.5% 0 1%<03% < 1.8%

Intenor Finish 45.5% < 49.8% < 54.1% 41.8% < 46.0% < 50.3% 2.4% < 4.2% < 7.3% 0.0% < 0.0%. < 1.1%

Foundation 161,300 < 165,600 < 169,800 10.700 < 14,600 < 18,400 600 < 1,700 < 4,900 600 < 1,700 < 4,900

Foundation-to-Walls 161,800 < 166,000 < 170,200 9,900 < 13,600 < 17,300 900 < 2,800 < 6,500 400 < 1,100 < 4, 100

Walls 172,600 < 180,000 < 181.700 1,600 < 3,500 < 7,500 0<0 < 1,700 0<0< 1,700

Roof 178,400 < 182,400 < 182,700 300 < 1,100 < 4,000 0<0< 1,700 0<0< 1,700

Exterior Finish 85,600 < 93,000 < 100,300 77.200 < 84,600 < 91,900 2,900 < 5,400 < 10,100 100 < 600 < 3,300

Intenor Firush 83,500 < 91,400 < 99,300 76,600 < 84,500 < 92,400 4,300 < 7,600 < 13,400 0<0<2,100

359
Table 3

Construction Characteristics

of Sampled Single-Family Detached Homes

Component Hurricane Andrew 1992

No. of Stories 80% One 2% Other

18% Two

Roof Construction 81% Gable 6% Other

13% Hip

Wall Construction 96% Masonry

4% Wood Frame

Foundation Type 100% Slab

Siding Material 94% Stucco 6% Other

Roofing Material 73% Comp. Shingle

18% Tile 9% Other

Interior Finish Primarily Gypsum Board

Table 4

Percentage of Sampled Single-Family Detached Homes

With 'Moderate' or 'High' Damage Ratings

Component Hurricane Andrew 1992

Roof Sheathing 24% (64%)'

Walls 2%

Foundation 0%

Roof Covering 77% (99%)'

Interior Finish (water damage) 85%

'Percentage in parenthesis includes "low" damage rating and, therefore, corresponds

to homes with one or more sheathing panels lost or any form of roof covering

damage. Other values indicate moderate or high damage ratings, including roof blow-off or similar serious failures (i.e., collapse).

Table 5

Summary of Exterior Finish Wall Damage

to the Sub-Sample of 49 Homes

Damage Rating Percent of Survey

NONE 55.1 LOW 40.8

MODERATE 4.1 HIGH 0.0

360
Table 6

Summary of Strong Ground Motion Stations

in Close Proximity to the Sampled Homes

Station Name Latitude Longitude 0.2 sec Spectral

Response

Acceleration

0.2 sec Return

Period (yrs)

1.0 sec Spectral

Response

Acceleration

(%g)

LO sec Return

Period (yrs)

Lentury Lity - LALL iNtn J4.Uo4 1 16.41 / 06V0 43% 432

LA - ULLA Urounds 1 1 B.4jy VUvo 1 C\ 1 ^ AG7 195

T A - Hn! 1 V \A/nr>H ^tr^raof* T nt 34.090 1 1 8 339 100% Z. O 1 '-r\J /(J Tarzana - Cedar Hill 34.160 118.534 270% 13,619 79% 1,813

Arleta - Nordhoff Station 34.236 118.439 81% 169 53% 502

Pacoima - Kagei & Canyon 34.288 118.375 76% 116 53% 326

Malibu - Point Dume 34.013 118.800 26% 139 10% 209

Moorrpark - Fire Station 34.288 118.881 63% 108 23% 99

Newhall - Fire Station 34.387 118.530 140% 268 117% 3.021

\

Table 7

Summary of Homes Surveyed on Mecca Avenue

Address

Solid Wall

Ratio, P

No. of Wall

Offsets

Closest

Station

Damage

Rating

#1 Mecca Ave 0.349 2 Tarzana' None

#2 Mecca Ave 0.389 0 Tarzana' Low

#3 Mecca Ave 0.455 0 Tarzana' Moderate

|

'The strong motion readings at the Tarzana ground station were among the largest recorded during the Northridge

Earthquake.

361

American Lifelines Alliance

Reducing Risks to Utility and Transportation Systems from Natural Hazards

by

Stuart Nishenko', Tim Sheckler', William U. Savage^ and Douglas Honegger^

ABSTRACT

The Federal Emergency Management Agency

(FEMA) and the American Society of Civil

Engineers (ASCE) have entered into a cooperative agreement to establish the

American Lifelines Alliance (ALA) publicprivate partnership project, with the goal of

reducing risks to lifelines (utility and

transportation systems) from natural hazards.

ALA's objective is to facilitate the development

and adoption of national consensus design and

retrofit guidelines and standards that, when

implemented by lifeline owners and operators,

will systematically improve the performance of

lifelines to acceptable levels in natural hazard

events. ALA is working with partners on existing nonconsensus guidelines, standards, and industry

practices, synthesizing and improving them, as

appropriate, and shepherding the resulting

documents through a formal consensus process. Each project involves (1) a Standards

Developing Organization (SDO) for consensus

development; (2) representatives from the

appropriate sector of the utility or transportation

community; and (3) representatives from the

relevant manufacturing and technical

communities.

ALA is using a general, well-established

procedure for assessing the performance of

utility and transportation systems subjected to natural hazards, and identifying actions to reduce their risk. When ALA-sponsored

consensus guidelines and standards are available

to implement this procedure and are applied to a

specific utility or transportation system, the

likely performance of the system in a natural

hazard occurrence is quantified, and guidance

for possible mitigation actions is provided.

ALA's initial activities are focused on

converting well-established practices within the

utility and transportation industries into

nationally applicable consensus guidelines and

standards. ALA is seeking sponsoring

organizations and stakeholders to co-fund and

provide in-kind support to mutually beneficial

efforts. In some cases, existing practices are not

sufficient to be put through a formal consensus

process. In these situations, the role of ALA

may include funding studies in order to improve

or extend practices to the point where national consensus is achievable. ALA is also

establishing extensive outreach and

implementation activities.

KEYWORDS: American Lifelines Alliance,

lifelines, utilities, transportation systems,

' Federal Emergency Management Agency,

500 C Street SW, Washington DC 20472

^ PI/ALA Fellow, Federal Emergency Management

Agency, Washington D. C, 20472 and Pacific Gas and

Electric Company, San Francisco, CA 94105

^ 2690 Shetland Place, Arroyo Grande, CA 93420

363
natural hazards, risk assessment, performance

guidelines

1.0 INTRODUCTION

The Federal Emergency Management Agency

(FEMA) and the American Society of Civil

Engineers (ASCE) have entered into a

cooperative agreement to establish the

American Lifelines Alliance (ALA) publicprivate partnership project. The vision of the

project is to reduce risks to lifelines (utility and

transportation systems) from natural hazards.

ALA's objective is to facilitate the creation,

adoption, and implementation of American

National Standards Institute (ANSI)-approved

national consensus design and retrofit guidelines

and standards. These documents, when

implemented by lifeline owners and operators,

will systematically improve the performance of

utility and transportation systems to acceptable

levels in natural hazard events. Formation of

the ALA is in direct response to needs identified over 10 years ago, and is an outgrowth of a request by Congress in the 1990 Reauthorization

of the National Earthquake Hazards Reduction

Program (NEHRP).

The process that is being developed and

implemented by ALA addresses considerations

that are unique to or particularly important to transportation and utility systems.

• Each lifeline system is unique in its development history, and each system

typically occupies a region that has

geographically varying exposure to natural

hazards. The evaluation of performance of

the system must be able to accommodate the

specific components of the system, which

may have been designed according to different criteria at different times, and are exposed to site-specific natural hazards.

• Most lifeline systems include networks or

multiple components that provide

redundancy and resilience when subjected

to damage or disruption. As a consequence

of this inherent functionality, a given

lifeline can tolerate some level of damage

while providing adequate performance as

far as the users of the lifeline are concerned.

The analysis of risk due to natural hazards

must incorporate consideration of

functionality as experienced by the user of

the lifeline system.

• The consensus guidance products that result from ALA activities are being prepared to be used on a voluntary basis by owners and

operators of lifelines; but they could also be

adopted by lifeline regulators to provide

suggested or required guidance for achieving acceptable performance in response to natural hazards. Acceptable

performance criteria must be explicitly

established by the appropriate stakeholders,

including the lifeline agency, its customers

or users, and cognizant regulators.

Acceptable performance criteria must

include considerations of redundancy, and

should preferably be characterized in probabilistic terms.

This paper describes the approach to the

development of lifeline guidelines and standards

being used in ALA. A general procedure

commonly being used for natural-hazards risk management serves as a template for ALA

product development for all lifelines. At the

present time, guideline elements are being

developed in several lifeline areas.

Systems suitable for the ALA process include:

Electric power transmission and distribution

Natural gas transmission and distribution

Potable water conveyance and distribution

Wastewater transportation and processing

Oil and liquid fuel handling, transport, and

storage

Highways

Railroads

Telecommunications

2.0 ROLE AND SCOPE OF ALA

Under the guidance of a working group, ALA

has identified lifeline systems for which non- consensus guidelines, standards, and industry

practices exist. ALA is taking these existing

documents, synthesizing and improving them,

364
as appropriate, and shepherding the resuhing

documents through a formal consensus process

to develop nationally applicable guidelines.

Each project involves a Standards Developing

Organization (SDO) for consensus development, representatives from the

appropriate sector of the utility or transportation

community, and representatives from the

relevant manufacturing and technical

communities. Placing each consensus document

in an SDO will ensure that the document will be

reviewed and updated on a regular basis, so that

it remains current with respect to new hazards

data, updated system performance evaluation

procedures, and changes in society's acceptable

levels of risk. ALA is working with a wide

range of sponsoring organizations and

stakeholders to co-fund and provide in-kind

support for specific activities. ALA project

working group members and FEMA personnel

provide coordination and oversight of these

activities.

In some cases, existing practices for natural

hazards risk management are not developed

enough to be put through a formal consensus

process. For example, databases for ice loading

hazards have been developed for only the

eastern half of the U. S., and procedures for

evaluating the performance of water conveyance

pipelines have not been comprehensively

assembled for the national range of pipe types,

soil conditions, levels of ground shaking, and

levels of permanent ground deformation. In these and similar situations, the role of ALA

includes funding or otherwise encouraging

studies and projects that will improve or extend

practices to the point where national consensus

is achievable.

2. 1 General Risk Analysis Procedure

ALA is using a general procedure for assessing

tiie performance of utility and transportation

systems subjected to natural hazards, and

identifying actions to reduce their risk. The

following six steps describe the procedure.

• Identify and quantify the natural hazard

• Identify and quantify the expected damage

states of components within the system

• Assess the damage states and functionality

of the system components when affected by

a natural hazard

• Evaluate system functionality

• Identify and evaluate impacts of reductions

of lifeline system functionality and financial or other losses

• Provide guidance to the system owner and

others

When ALA-sponsored national consensus

standards and guidelines are applied to a

specific utility or transportation system, the

likely performance of the system in a natural

hazard occurrence is quantified, and guidance

for possible mitigation actions is provided.

2.2 Implementation of ALA Guidance

ALA is also undertaking programs to encourage

the implementation of natural hazards risk

reduction measures by system owners and

others. These programs consider the many

means by which natural hazard risks can be

reduced within the diverse types and sizes of

lifeline systems, and address owner concerns over budgeting, coordination with other internal

programs, management support for risk

reduction, and regulatory issues. Also, those

who are currently implementing natural hazards

risk-reduction programs are being asked to identify particularly useful existing guidelines

and standards, as well as barriers to their

implementation. ALA is coordinating these

efforts with other public/private partnerships

such as FEMA's Project Impact.

2.3 -Status of ALA Activities

Since its inception in 1998, ALA has developed

and begun to carry out the program described

above. FEMA funding has been critical in supporting these initial ALA activities: • An increasingly broad community—from

lifeline engineers to regulators to lifeline owners and operators—is being informed of

ALA plans and activities. A group of more

than 200 Corresponding Advisors has been

established to provide input and feedback

365
on ALA. plans.

Several efforts have been or are being

funded by the ALA project to take

advantage of well-established industry

practices.

1. An engineering firm is preparing a comprehensive procedure for

establishing earthquake vulnerability

relationships for potable water

conveyance pipelines and related

facilities such canals, tunnels, storage

tanks, valves, and control

instrumentation. The 'vulnerability

relationships determine the probability

of specified damage states as a function

of pipe or facility characteristics, and

earthquake hazard characteristics

including soil conditions, peak ground

motions, permanent ground

deformation, and other relevant

parameters. This procedure will be

incorporated in a subsequent

development of earthquake

performance-based design guidance.

2. An existing standards committee task

group is being provided with logistical

support to complete development of

earthquake performance criteria for gas

and liquid fuels transmission pipelines.

3. An existing ice-loading hazard map that was" developed for several regions in the

United States is being extended to cover

the entire U. S. east of the Rocky

Mountains. This map will be available

to be incorporated as an appendix of the

ASCE 7 Standard, Minimum Design

Loads for Buildings and Other

Structures. FEMA, ASCE, and Pacific Gas and Electric

Company (PG&E) have established and co- fund a fellowship position at FEMA. The

fellowship recipient (One of the authors of

this paper) is spending one-quarter time for a year working at FEMA to bring utility

expertise into FEMA's Mitigation

Directorate in particular support of ALArelated activities. Electric power earthquake risk is one topic

of the multi-participant, applied earthquake

research program established in 1996 by

PG&E and the Pacific Earthquake

Engineering Research Center (PEER) at the

University of California at Berkeley. The

California Energy Commission, the

California Department of Transportation,

and other utilities are also providing

research support. Utility-directed studies of

the seismic vulnerability of high-voltage

substation equipment, seismic interactions

that potentially damage interconnected

substation equipment, and predictive

functionality models for electric power

system performance in earthquake scenarios

are being conducted, with the goal of

incorporating the results into an ALAdeveloped earthquake design guideline.

4.0 CONCLUSIONS

The American Lifeline Alliance has been

established as a major initiative supported by

FEMA with the goal of establishing a national consensus process to reduce risks to lifelines

(utility and transportation systems) from natural

hazards. The ALA effort draws on the

experience and expertise of lifelines personnel,

consultants, academic researchers, and others to build consensus agreements on practical and

effective means to accomplish this goal. With

the involvement of many organizations,

agencies, and individuals over the next several

years, ALA is expected to oversee the

development and implementation of consensus

guidelines and standards that, over time, will

achieve a significant improvement in the

performance of lifelines affected by natural

hazards.

'

ACKNOWLEDGEMENTS

The authors acknowledge the input to this paper

derived from documents prepared by the ALA

Working Group. Thomas McLane provided

additional ideas and reviews. The support of

FEMA, ASCE, and PG&E is gratefully

acknowledged.

366
Theme 5

Public Health

367
I

I

I

368
Public Health Assessments in Post-disaster Settings:

Recommendations for a Multidisciplinary Approach

by

Josephine Malilay^

ABSTRACT

Rapid assessments of needs and health status have

been conducted by the U.S. Centers for Disease

Control and Prevention (CDC) in natural disaster

settings for gathering information about the status of affected populations during emergencies. A

review of 8 such assessments (6 from hurricanes,

1 from an ice storm, and 1 from an earthquake)

examines current methods and applications, and

describes the use of results by policy makers so assessments in post-disaster settings can be

improved.

Seven of the 8 (88%) were performed during or

after the third day post-impact (range: 1-70 days,

median: 7 days). All addressed demographics,

morbidity, and water; 7 concerned food,

sanitation, and transportation; and 6 queried access to medical care and electricity. All 8 provided estimates of the proportions of needs; none, however, estimated the extent of needs for

the affected population. Of the 8, 5 confirmed a

policy decision, such as rescheduling food

delivery to an earlier date.

Increasing application of health assessments

provides impetus for improving current methodologies; standardizing collection

instruments; involving other sectors in emergency

relief; and ensuring useful information for decision makers.

KEYWORDS: needs assessments, natural

disasters, multidisciplinary assessments

1.0 INTRODUCTION

Obtaining timely, accurate, and reliable information

about the needs and health status of affected

populations presents a challenge to health workers

who must respond appropriately and effectively to catastrophic emergencies and disaster situations. In acute settings such as post-natural disastersdefined as ecologic disruptions exceeding the local capacity to respond and calling for outside

assistance— information about changing events and

needs becomes critical as large numbers of people

may become injured or ill or are killed, populations

displaced, living conditions overcrowded, usual

sanitation and hygiene compromised, and normal

public health programs interrupted or ceased. Such

information becomes requisite for emergency

officials to make appropriate requests for urgent

medical care or relief assistance.

In recent years, rapid health assessmentsconducted by program managers to monitor and

improve the performance of health-care systemshave become the modi operandi for gathering

information about the status of an affected

population, particularly during emergency

response. Known also as rapid needs assessments

or rapid epidemiologic assessments, these

assessments represent the first line of

epidemiologic response activities in the aftermath

of disasters, with results providing the basis for

directing relief efforts. Although earlier methods

have been largely ad hoc and fragmentary,

subsequent procedures use epidemiologic study

designs and sampling approaches.

Centers for Disease Control and Prevention, National Center for Environmental Health,

1600 Clifton Road, NE (Mailstop E-23), Atlanta, Georgia 30333 USA

369
The cluster sampling approach for assessing postdisaster needs has gained increasing attention by

health subsectors providing initial response

activities, such as search and rescue, disaster

medicine, and emergency relief. After repeated

applications in actual field settings, the design has

been modified to yield reasonably accurate results that may be extrapolated to estimate the needs and

the magnitude of those needs in an affected

community.

Assessments of health needs have been conducted

systematically in post-natural disaster settings by

the U.S. Centers for Disease Control and

Prevention (CDC), a public health agency tasked

to address the health consequences of disasters. The specific contents of these assessments are usually at the discretion of the investigators and

the local health agency who conduct the actual

assessments in the field setting. Usually, with the

advice and collaboration of local officials or sponsoring organizations, rapid needs assessments have addressed 1) general needs that may be

common among affected populations (e.g., access

to transportation and prescription medications),

and 2) specific needs that may be germane to local conditions (e.g., access to particular

communications media such as a radio station), or

to the nature of a particular disaster event (e.g., use of gasoline-powered generators in an ice storm.)

Because the results of rapid health assessments

greatly influence the nature of relief activities, a review of past applications allows us to: 1) ascertain strengths and limitations of the

assessments; 2) examine the methods employed;

and 3) review the usefulness of information

generated and the subsequent use of this information by policy makers, so that lessons may

be learned to improve health needs assessments in post-disaster settings. This review compares

assessments for similarities and differences: 1) across disaster types; 2) within similar disasters;

3) by timing when the assessments are conducted;

and 4) in U.S.-based versus international settings.

It further identifies decision making actions that have resulted from information gathered, and

finally, suggests direction for ftiture assessments of

the needs and health status of affected

communities.

2.0 METHODS

Post-disaster assessments that were reported in CDC's Morbidity and Mortality Weekly Report, a weekly publication of current publication of current domestic and international public health issues,

from 1980 through 1999 were reviewed because

they applied a systematic methodology in data

collection. Selected assessments were compared

descriptively for the timing in which they were

conducted, their content areas, sampling

methodology, and results.

3.0 RESULTS

Thirteen needs assessments from 6 published

reports were identified for review. Of the 13, 3 were from Hurricane Andrew in Florida and 2 were

from Hurricane Andrew's subsequent strike on

Louisiana, USA; 4 were from Hurricane Marilyn in the U.S. Virgin Islands, and 1 was from Hurricane

Opal in the Florida panhandle; 1 was from

Hurricane Georges in the Dominican Republic; 1 was from a severe ice storm in Maine; and 1 was

from a major earthquake in Turkey. Five

assessments from 2 separate hurricane eventsMarilyn (4 assessments) and Opal (1 assessment)-- were described in 1 report. Four assessments from

Hurricane Andrew and its effects in 2 states, where

assessments were repeated within 7 days of each

other, were described in 1 report; a third

assessment in one of those states was performed 2

months later and reported separately. The

remaining 3 assessments—from an ice storm,

Hurricane Georges, and an earthquake—were

reported individually. The earliest assessment was

published in 1992, the latest in 1999.

Because repeated assessments conducted within

days of each other addressed the same basic items

for detecting changes in needs, this analysis

focuses on information from the first assessment,

with the exception of the third and independent

assessment related to Hurricane Andrew conducted

370
in Florida 2 months after the event. Therefore, 8

assessments are represented in this analysis. The

Seven (88%) of the 8 assessments concerned

climatic disasters, which were concentrated in U.S. states bordering the Gulf of Mexico and a

U.S. territory and an island nation in the

Caribbean Sea. Only 1 assessment addressed a geologic disaster, a major earthquake in Turkey.

Seven of the 8 assessments were performed

during or after the third day post-impact (range: 1 day to 70 days, median: 7 days). Assessments in the 2 foreign countries were performed at least 60

days after the disaster event.

3.1 Demographics, Food, Health Care, and

Lifeline Systems

All 8 assessments addressed demographics

(number and ages of household occupants),

morbidity (types and number of sick or injured),

and water supply. Seven queried residents about

sanitation, food supply, and transportation; 6

inquired about access to medical care and

electricity. The contents of the assessments are presented in Table 2.

For climatic disasters, the most frequent issues

concerned demographics, morbidity, medical care, food, water, and sanitation. For the survey

conducted 2 months after Hurricane Andrew,

additional in-depth questions were raised about

barriers to adequate care, health insurance,

unemployment compensation, public health

programs, primary health care including

preventive services, nonspecific environmental

problems, mental health status, and evacuation

behavior. Vulnerable subgroups in the population

such as young children, the elderly, pregnant and

lactating women, and migrant workers who may

have marginal access to relief services were

identified in the assessment after Hurricane

Georges in the Dominican Republic. Similar to the 2-month survey conducted after Hurricane

Andrew, questions after the ice storm were

specific for homes without restored electrical power, including sources for generating alternate power and the presence of working carbon

Finally, we determined whether findings resulted

in actions taken by decision makers such as health

characteristics of the study assessments are presented in Table '1

.

monoxide detectors.

Although topics for the single geologic event, an

earthquake, were similar to those for hurricanes,

the striking difference was the question related to

disaster-related deaths in the household. No other

disaster event presented a question about mortality.

Assessments conducted domestically and abroad were similar in content.

3.2 Populations at Risk

Because the primary intention of a needs

assessment is to describe the scope of the needs of

affected populations and ideally, the magnitude of

those needs, the reports were examined for

references to populations at risk. Five of the 8

assessments provided estimates of the pre-event

population, usually from the census; 3, assessments

from Hurricane Andrew in Florida and Louisiana

and Hurricane Opal in the Florida panhandle, did

not. All 8 assessments estimated needs based on

proportions of needs, using population estimates

derived from the demographics portion of the

survey. None, however, extrapolated the

magnitude of needs to estimate the extent of needs

for populations at risk. For example, 52% of

respondents from the example of Hurricane

Marilyn had access to electrical power, indicating

that 48% were without electricity at the time of the assessment. The proportion may represent 480 of

1,000 people, 4,800 of 10,000, or 48,000 of

100,000. Without knowing the population at risk

and given that census estimates and electrical power records were unavailable for immediate

decision-making, it is difficult to determine the

magnitude of the need for electricity at the time of

the survey, e.g., whether the utility needed to be

restored for the households of 480 or 48,000

persons.

3 .3 Actions Taken By Decision Makers

authorities and emergency management officials,

and whether such actions were documented in the

371
report. Five of the 8 assessments confirmed

actions in response to the results, such as rescheduling food delivery earlier than planned

and discontinuing plans for house to house

outreach when most residents had access to a

particular radio station. One assessment, from the

ice storm, demonstrated that although the lack of

electricity was extensive and continued into a second week, most residents were able to meet

their needs for food, water, warmth, and

sanitation. Two assessments, from Hurricane

Opal in Florida and the earthquake in Turkey, did not document any actions taken by decision

makers, although results were conveyed to local health authorities.

4.0 DISCUSSION

The primary intention of needs assessment is to provide reasonably accurate and reliable estimates

of needs with relatively quick turnaround so that appropriate actions for relief can be taken.

However, all but 1 of the assessments actually were conducted > 3 days following the event. Because these data suggest that logistical

constraints may delay the arrival of an assessment

team in less than 3 days, perhaps local authorities or neighboring institutions or persons should be

designated by local authorities to perform

assessments in the event of a major disaster. The

relative simplicity of the survey instrument and an

orientation to survey techniques, including the

modified cluster sampling method, could be

facilitated in train-the-trainer programs for

emergency management, particularly for the

health sector. By the same token, a needs

assessment may not be realistic or practical to be

conducted locally in the first 72 hours postdisaster as people are engaged in search and

rescue activities, first aid, and burial of the dead.

Interestingly, all assessments used proportions

based on a denominator composed of the sample

population to measure the scope, or range, of

needs in the surveyed community. Although this Assessments of needs and health status raise other

post-disaster interests that may transcend the

public health response community, particularly

among professions related to prehospital

is a reasonable first step, the magnitude, or size or

extent of those needs should also be estimated to

reflect the needs of the entire affected community.

This may be accomplished by obtaining estimates

of the number of people affected by the disaster

from sources such as the census or local authorities. By determining the magnitude of needs of the

affected population, estimates can then be derived

for response actions, such as the quantity of

resources to restore electricity to a certain number

of households.

Basic categories of needs and health status were

uniform for the most part, particularly with regard

to demographic composition of households; health

status; medical care; food; water and sanitation; access to lifeline systems; housing damage; and

shelter requirements. After 8 applications of the

assessments, standardizing the collection, analysis,

and interpretation of these elements may be a next

step. A recent worldwide attempt, using the

experience of nongovernmental organizations

(NGOs) in humanitarian assistance, is the Sphere

Project, This conglomeration of NGOs produced a

charter promulgated by more than 228

organizations from more than 60 countries and

identified minimum standards for disaster

assistance in water supply and sanitation, nutrition,

food aid, shelter and site planning, and health

services. Perhaps a similar effort should be

spearheaded for performing needs assessments in disaster settings, as has been recommended for

situations of complex emergencies. Assessments

are performed by many groups that inundate a

stricken area to provide aid. The methods for the assessments are frequently undocumented, may not be systematic, and often present difficulty in interpretation and generalization to the entire

disaster-affected area. Standardization of the content of needs assessments would facilitate identification of needs and delivery of appropriate

services and resources in the aftermath of a

disaster.

emergency and disaster medicine and engineering

for lifeline systems. Among responders from

emergency and disaster medicine, information is needed as to what may be considered useful in

372
assessments conducted within 3 days, 3 to 10

days, and more than 10 days after a major disaster. For example, within 3 days of arrival, appropriate

activities that may be incorporated in a needs

assessment include evaluating patient volumes at emergency medical-care facilities; assessing

pharmaceutical stockpiles during the emergency

period; and projecting the use of stockpiles so that additional supplies may be requested in anticipation of the volume for ftiture demand and

use. Medical outreach teams have been proposed,

but the reality of their arrival immediately after a

disaster event raises questions about their usefulness during the emergency period. One

solution may be to train local personnel to perform

such assessments.

Lifeline systems, damage to pre-disaster housing,

and types of shelter were addressed in most of the assessments. Usually not directly linked to public

health, these topics are nonetheless associated

with human health and well-being after a disaster event. For example, a continual supply of

electrical power is needed for persons on home

health care, storage of medicines requiring

refrigeration, and heat and warmth for safe shelter. Almost immediately after a disaster, the integrity

of lifeline systems is examined in field assessments, similar to those performed for public

health. Including the disciplines that perform

these activities in needs assessments for public

health may facilitate efficiency, promote the

sharing of mutually beneficial information, and

consolidate appropriate resources for response.

This review included only assessments performed

by CDC and reported in the Morbidity and

Mortality Weekly Report. Many other assessments

are performed in post-disaster settings by local governments, NGOs, international organizations,

and military sectors from donor countries.

Nevertheless, this review suggests one trend in the

increasing frequency of applying post-disaster

needs assessments, consistent use of specific

issues, and ways in which the methods and the

performance of such assessments can be modified

to improve overall delivery of assessments in disaster field settings.

5.0 CONCLUSIONS

Although an element of urgency was associated

with them, needs assessments in this review were

typically conducted within 1 week after disasters

occurred. However, this delay did not obviate their

utility in the field setting depending on the

objectives of the assessments and changes in needs

with time. For assessments that may be performed

within 72 hours of a disaster, acute needs may be

related to immediate response, such as emergency

needs, use of emergency services, and

identification of appropriate emergency medical

personnel. Additional assessments from this time

have yet to be conducted for full evaluation. For

assessments performed 3-10 days post-disaster,

issues focused on demographics, health status, food

and water supply, and restoration of utilities and

services. Most of the assessments in this review were conducted during this time frame, and may be

indicative of the reality of time and logistical

constraints when responding to disaster events.

Finally, 3 assessments also were conducted more

than 1 month after the event. Categories of interest expanded to include others such as barriers to adequate care, mental health status, evacuation and

storm preparation behaviors, continuing food,

water, and sanitation needs, warmth, and shelter types.

Since 1992, needs assessments are increasingly

being applied in the field setting. The widespread

use of these techniques for disaster response

provides an impetus to improve current methodologies; standardize collection, analysis,

and presentation techniques; involve other

disciplines who are also working to achieve the same objectives; communicate results clearly to decision makers; ensure that information is usefiil; and finally, document actions taken by decision

makers in response to the results.

6.0 REFERENCES

Anker M. Epidemiological and statistical methods

for rapid health assessment: introduction. World

health statistics quarterly 1991;44:94-97.

Balthazar JC. The potential of the case-control

373
method for rapid epidemiological assessment. World health statistics quarterly 1991;44(3):140-

144.

Boss LP, Toole MJ, Yip R. Assessments of

Centers for Disease Control and Prevention. Rapid

health needs assessment following Hurricane

Andrew — Florida and Louisiana, 1992. Morbidity

and Mortality Weekly Report 1992;41(37):685-

688.

Centers for Disease Control and Prevention.

Comprehensive assessment of health needs 2 months after Hurricane Andrew — Dade County,

Florida, 1992. Morbidity and Mortality Weekly

Report 1993;42(22):434-437.

Centers for Disease Control and Prevention.

Surveillance for injuries and illnesses and rapid

health-needs assessment following Hurricanes

Marilyn and Opal, September-October 1995.

Morbidity and Mortality Weekly Report

1996;45(04):81-85.

Centers for Disease Control and Prevention.

Community needs assessment and morbidity

surveillance following an ice storm — Maine,

1998. Morbidity and Mortality Weekly Report

1998;47(17):351-354.

Centers for Disease Control and Prevention.

Needs assessment following Hurricane Georges —

Dominican Republic, 1998. Morbidity and

Mortality Weekly Report 1999;48(05):93-95.

Centers for Disease Control and Prevention.

Community needs assessment and morbidity

surveillance following an earthquake — Turkey,

August 1999. Morbidity and Mortality Weekly

Report 1999;48(50): 1147-1 150.

Guha-Sapir D. Rapid assessment of health needs

in mass emergencies: review of current concepts

and methods. World health statistics quarterly

1991;44(3):171-181.

Gunn SWA. Multilingual dictionary of disaster

medicine and international relief. Dordrecht, the

Netherlands: Kluwer Academic Publishers, 1990.

mortality, morbidity, and nutritional status in Somalia during the 1991-1992 Famine. Journal of

the American Medical Association 1994;272:371-

376.

Hlady G, Quenemoen LE, Armenia-Cope RR, et al. Use of a modified cluster sampling methods to perform rapid needs assessment after Hurricane

Andrew. Annals of Emergency Medicine

1994;23:719-725.

Leonard RB, Spangler HM, Stringer LW. Medical

outreach after Hurricane Marilyn. Prehospital and

Disaster Medicine 1 997; 1 2(3): 1 89- 1 94.

Lillibridge SR, Noji EK, Burkle FM. Disaster assessment: the emergency health evaluation of a population affected by a disaster. Annals of

Emergency Medicine 1 993 ;22: 1 7 1 5- 1 720.

Malilay J, Flanders WD, Brogan D. A modified

cluster-sampling method for post-disaster rapid

assessment of needs. Bulletin of the World Health

Organization 1996;74:399-405.

Pan American Health Organization. Epidemiologic

surveillance after natural disaster. Scientific

Publication No. 420. Washington, D.C.: PAHO,

1982.

Smith GS. Development of rapid epidemiological

assessments methods to evaluate health status and

delivery of health services. International Journal of

Epidemiology 1989;18:S2-S15.

The Sphere Project. Humanitarian charter and

minimum standards in disaster response. Oxford,

United Kingdom: Oxford Publishing, 2000.

374
Table 1 . Characteristics of 8 needs assessments from the Centers for'Disease Control and Prevention,

Morbidity and Mortality Weekly Report, 1980-2000

Characteristic No. of assessments

Disaster type

Climatic (hurricane, ice storm) 7

Geologic (earthquake) 1

Timing of assessment < 3 days post-event 1 3-10 days 4 > 1 0 days 3

Number of assessments of same event

Hurricane Marilyn, U.S. Virgin Islands 4

Hurricane Andrew, Florida 3 Hurricane Andrew, Louisiana 2

Sampling method

Systematic 1 Cluster

Modified 1 -stage 6

2-stage 1

Estimate of population at risk in disaster zone addressed

^es 7

No 1

Type of information provided

Proportion of needs (%) 8

Extent of needs extrapolated for population at risk 0

Actions taken by decision makers

Yes, documented 5 No, unneeded 1 Unknown 2

375
Table 2. Contents of assessments and time frame in which assessments were conducted

No. of assessments performed within days of event

Topic < 3 days

n=l

3-10 days

n=4

> 1 0 days

n=3

Total

Demographics

No. and ages of

household occupants 1 4 3 8

Vulnerable subgroups < 2 years, > 65 years, pregnant female

1 2

2

32

Health status

Morbidity

types

no. of sick/injured

1

1

4

4

33

8

8

Mortality (no. of deaths) — — 1 1

Medical care (Needs and access to prescription medicine) — 4 2 6

Food supply

(Availability) 4 3 7

Water supply

(Availability, source, and

treatment) 1 4 3 8

Sanitation

(Access to working toilet) 1 4 2 7

Housing damage — 2 2 4

Shelter type — 2 2 4

Access to lifeline systems

Electricity

(Status and sources) 1 4 1 6

Communications 1 4 5

Transportation 1 4 2 7

376
ASSESSMENT MODELS FOR EARTHQUAKE FATALITIES AND AN

INTERPRETATION OF THE DEATHS IN THE 1999

KOCAELI, TURKEY, EARTHQUAKE"

by

Keishi SmONO"

ABSTRACT

Two models for estimating earthquake fatalities were developed on the basis of existing

knowledge in seismology, earthquake

engineering, and natural disaster sciences.

Those were designed so that disasters in the wide range of built environment are coherently

examined with a limited range of input

information. The performance accuracy of the models was examined and confirmed

acceptable. One of the two models was applied

to the discussion on the nature of the fatalities

in the Kocaeli earthquake. It was pointed out

that the enhancement of structural quality of

multi-story apartment buildings is crucial toward the reduction of human losses in future earthquakes.

KEY WORDS: Earthquake Disaster

Damage Estimation

Mortality

Kocaeli Earthquake

1. INTRODUCTION

It is easy to understand that earthquake

casualties depend on: 1) earthquake, 2)

building collapse, and 3) affected population.

Even non-professional people can tell that more victims are killed if an earthquake is more violent; if more buildings collapse; and if more people stay in a building.

Quantitative evaluation of earthquake fatalities, however, has seldom been attempted despite its significance in discussion toward the reduction

of human casualties. Though not a few

methods for correlating building damage to

human losses were developed, those

applicability was rather limited. Those were

useful only in areas having the same built environment, which is characterized mainly by

the construction type of local buildings.

Mass casualty in earthquakes is a serious issue

from the viewpoint of global disaster

management. With relation to the issue, it is crucial to have an estimation model that can

coherently handle disasters in areas with

different built environment.

In the first half of this paper, two models for estimating earthquake fatalities, which require

only a limited range of input information, but

applicable to disasters in the wide range of

built environment were developed. In the last half, one of the two models was applied to the discussion on the nature of the fatality in the M7.4 Kocaeli, Turkey, earthquake of August

17, 1999.

2. TWO MODELS

Two models as follows were developed in this study:

1) Distribution model

2) Point model.

1) This is a revision of a paper presented in Third Japan-Turkey Workshop on

Earthquake Engineering, February 21-25,

2000, Istanbul, Turkey.

2) Department of Civil Engineering, Nagaoka

College of Technology, Nagaoka, Niigata

940-8532 Japan

377
The most significant difference between the two models is found in the method for giving

seismic information. It was given by the spatial

distribution of seismic intensity in the

distribution model, while it was given by a

single parameter of the earthquake magnitude

in the point model. The point model did not

include the spatial variation of seismic intensity

in an explicit manner, but one of the

relationships used in the model included the

attenuation of seismic intensity.

In the point model, population density was

given as that for the entire affected area. Building information was given as dominant

construction types chosen from a nine-category

classification. The classification was done

based on the type of principal load-bearing

elements and wall material. Detailed

explanation is seen in the following section.

The point model is simple, because it was

given as a single formula. Only substitution of

three independent variables into the formula is required for deriving an estimate. A flaw in the point model lies in the exclusion of spatial

relationship, or distance, between an epicenter

and affected areas. However, this does not reduce the model's applicability, because most

earthquake disasters are attributed to earthquake faults located near affected areas.

In the distribution model, earthquake

information was given by the spatial

distribution of seismic intensity. The spatial distribution of population density can be

included if necessary. Building types were

assigned from the nine-category classification, as was done in the point model, and their spatial disfribution can be included if necessary. Numerical procedure in the

distribution model is more precise in comparison with that in the point model. The

numerical procedure in the distribution model,

which requires a small computer program, is a

little more complicated than that in the point

model.

Characteristics of the two estimation models

are compared in Table 1

.

Only the general ideas for constructing the models and the performance accuracy of the

models are described in this paper. Results of

case studies for verifying the models are seen

in Shiono et al. (1991).

3. DISTRIBUTION MODEL

(1) Structure

Figure 1 shows the flow of calculation in the

distribution model. Figure 2 is schematic

presentation of the numerical procedure in the

distribution model.

Input information was composed of two parts:

1) Earthquake information

2) Regional information.

Earthquake information includes:

1) Surface wave magnitude

2) Epicentral location.

Regional information consists of:

1) Population density

2) Building type (s) 3) Site condition in terms of increment in seismic intensity.

Geographical area in which the three regional

data assigned must be coincide with an

affected area. Spatial distribution of regional

data and a combination of dominant

construction types can be included if necessary. A nine-category classification representing

worldwide construction types consists of:

1) Field stone

2) Adobe

3) Cut stone

4) Brick

5) Wood frame with thin infill walls

) Wood frame with thick infill walls

7) Wood

8) Poor-quality reinforced concrete

9) Good-quality reinforced concrete.

378
Buildings of field stone and adobe correspond

to structures classified in Category A of the MSK seismic intensity scale. Buildings of cut

stone, brick, and wood frame with infill walls, or half timbered structures, are in Category B. Buildings of wood and good-quality reinforced concrete are in Category C. Buildings of

poor-quality reinforced concrete need further discussion to be classified in an appropriate

category.

Relationships, or knowledge, used in the

distribution model were as follows:

1) Epicentral seismic intensity versus magnitude

2) Attenuation of seismic intensity with

distance (Attenuation function)

3) Structural vulnerability of each

construction type versus seismic intensity

4) Fatality rate (deaths/population) in each

construction type versus collapse rate.

Distribution of seismic intensity was

determined from: 1) magnitude versus

epicentral intensity relationship, such as Kamik

(1965), and 2) an attenuation function, such as Chandra (1979). A simplified approximation of

the circular distribution of seismic intensity was employed in this study.

A vulnerability function, which is given as a

relationship between seismic intensity and

damage extent sustained by a single building, was defmed for each construction type, and a

set of nine collapse rate functions was

generated to represent the total of construction

types (Figure 3). Detailed explanation of the

procedure for deriving collapse rate functions

from vubierability functions is found in Shiono

etal. (1991).

Fatality rare functions were determined based on published survey results, such as Cobum et

al. (1989). A set of three functions shown in Figure 4 was used in this study.

(2) Performance

Data from numerous past earthquakes exist to

examine performance accuracy of the

estimation model. Data were collected from 16 earthquakes (1962-1986) listed in Table 3. In collecting data, any criteria for selecting events was used, so that a performance test can be

done with as many data as possible. However,

several earthquakes in which damage was

limited to sites distant from the epicentral

regions, such as the Rumanian earthquake of

1977 and the Mexican earthquake of 1985 were excluded. For these disasters, a simplified

approximation of the circular distribution of

seismic intensity, which was used in this development, is no longer appropriate.

Death tolls calculated in the distributed model were plotted in Figure 5 in comparison with

reported fatalities. A correlation coefficient of

0.84 was obtained between the estimated and

reported death tolls for the entire (16) events,

and that of 0.90 was obtained for the 13 events

in which more than 1,000 lives were lost.

4. POINT MODEL

A simplified method, where spatial variation of

seismic intensity was not explicitly included, was constructed and named the point model. A

relationship derived from the distribution model, which implicitly includes the

attenuation of seismic intensity, was used in the point model. In handling of the point

model, programming in computers was not

required. Only substitution of three variables,

which respectively represent earthquake,

building, and demographic characteristics of a

disaster, into a single formula was necessary.

Input data to the point model were:

1) Surface wave magnitude

2) Dominant construction type (s) of local buildings

3) Population density (whole disaster area).

Among the three variables, magnitude was

chosen as the principal variable, and the

relationships between magnitude and fatalities was analyzed. Other two variables representing

379
buildings and population were used as

correction factors.

Figure 6 shows the relationship between

magnitude and crude death tolls for the earthquake disasters listed in Table 3. Positive

correlation between the two variables are seen, but the plots considerably scattered with a

correlation coefficient of 0.46. Logarithms of

death toll were used in the calculation of

correlation coefficients.

Figure 7 illustrates how magnitude correlates with death tolls when death tolls were adjusted on an assumption that population densities of

disaster areas are identical at 38 persons per

sq-km. A population density of 38 persons per sq-km was given as the global average of

population density. Scattering of the plots

reduced, and the correlation coefficient

increased up to 0.68.

The relationship between magnitude and deaths

calculated in the distribution model is shown in Figure 8. In the calculation for deriving this chart, population density was given at 100

persons per sq-km. Relationships were

essentially straight lines, and the lines were

nearly parallel to each other. In view of these

characteristics, ratios among the fatality rates in various construction types can be admitted

constant irrespective of earthquake magnitude.

The ratios were determined in Figure 8, and

the results are shovra in Table 3.

Figure 9 shows how magnitude correlates with

fatalities when those were normalized by

population density and corrected by

construction type. The ratios shown in Table 3 were used for including the difference in vulnerability and lethality among the

construction types. Scattering of the plots further reduced, and a correlation coefficient as high as 0.79 was obtained. A correlation

coefficient of 0.48 was obtained when only

correction by construction type took place,

suggesting the significance of simultaneous

inclusion of population and building

characteristics.

From the regression line derived in Figure 9

with the ratios shown Table 3, the following

formula was obtained:

D=(2.57 X 10^) X exp(2.93M) X (p/38)

2.2 (Field stone, Adobe)

1 .0 (Brick, Cut stone)

0.33 (Low-quality reinforced

concrete)

(Wood frame with thick

infill masonry walls)

(Wood frame with thin

infill masonry walls)

0.0017 (Wood)

0.12

0.026

J

where.

• • •(!)

D: deaths

M: magnitude (Ms)

p: population density of an affected area.

This equation can be applied to disaster areas having two or more dominant construction

types. In that case, population density is given

for each construction type, as total occupancies

in residential buildings of each construction

typeDeath tolls calculated in the two models were

plotted in Figure 10 in comparison with

reported tolls. Relative accuracy of the

estimates obtained in the two models can be

compared. Correlation coefficients between

reported and estimated death tolls were

obtained at 0.73 in the point model and at 0.84

in the distribution model, respectively. In the

regard of performance accuracy, the point

model is almost equal to the distribution model.

5. DEATHS IN THE KOCAELI

EARTHQUAKE

The Kocaeli earthquake (M7.4) occurred on August 17, 1999 in the northwestern part of

Turkey, and the death toll announced on September 13 was 15,480.

380
Post-event fatality estimation was done with

the point model (Eq. (1)).

Following assumptions were made in the

estimation:

1 ) Brick, half-timber (wood frame with thick

infill walls), and poor-quality reinforced concrete were the dominant construction

types in the affected area. 2) Brick and half-timber were structural types for single-family houses, while

reinforced concrete was that for

multi-story apartment buildings.

3) Ten families reside in an apartment

building.

With assumed regional data shown in Figure

1 1 and Table 4, Equation (1) gave an estimated

death toll of 15,380. The estimate was

acceptable, though further investigation on the

regional characteristics is indispensable.

A few additional cases were examined with

equation (1), and the result is shown in Figure

12.

The cases represent a probable scenario of

future transition in building type accompanying

with regional development. The scenario

includes, first, the rapid influx of population

into industrializing areas, among which is the

Kocaeli area severely damaged in the 1999

earthquake, and, second, increased demand for high-occupancy residential buildings, which are supplied as multi-story apartment buildings of

reinforced concrete.

In comparison with a population density of 227

persons per sq-km on Stage 0, which

represents the 1999 situation, that on Stage 3 is as high as 454 persons per sq-km. The increase

in population results in that in the number of

multi-story apartment buildings from 30% on

Stage 0 to 60% on Stage 3. Increased death

tolls on the later stages are attributed to the

structural vulnerability in high-occupancy

residential buildings of reinforced concrete.

6. CONCLUSIONS

Two models for estimating deaths in the wide

range of built environment affected by any

given earthquake were constructed. Those are simple and useful requiring a limited range of

input information consisting of earthquake,

building, and demographic data. Among the two models, one was named the distribution model, and the other the point model. The

distribution model includes the spatial variation

of input data, while the point model does not. Their performance accuracy was examined and

proved acceptable, though the point model was

sightly less accurate than the distribution model.

The point model was employed in the post-event estimation of the deaths in the 1999

Kocaeli earthquake. Remarkable agreement

between the estimate and the official

announcement was obtained, though the

building information, or predominant

construction types, was rather intuitive.

Death tolls in probable future circumstances of

the Kocaeli area due to hypothetical

recurrences of the 1999 earthquake were

estimated in the point model. Estimated death

tolls increase with the progress of regional

development accompanied by population influx and increased demand for high-occupancy

residential buildings. Structural quality

management of high-occupancy residential buildings, or multi-story apartment buildings of

reinforced concrete, is crucial toward life safety

in future earthquakes.

The author thanks Professor Y. Ohta of

Aichi-Shukutoku University and Professor H. Murakami of Yamaguchi University for their valuable contribution to this paper.

REFERENCES

Chandra, U. (1979): Attenuation of intensities

in the United States, Bulletin of the Seismological Society of America, Vol. 69,

No. 69, pp. 2003-2024

381
Cobum, A. W., A. Pomonis, and S. Sakai

(1989): Assessing strategies to reduce

fatalities in earthquakes. Proceedings of

International Workshop on Earthquake Injury

Epidemiology for Mitigation and Response,

July 10-12, 1989, Bahimore, Maryland, pp. PI07-P132

Karnik, V. (1965): Magnitude-intensity

relationship for European and Mediterranean

Table 1 : Comparison of input data between fatality estimation models

Model

Category of Data

Procedure

Earthquake Building Type Population Density

Point

Model

Magnitude Predominant

Types*

Average for the Whole Disaster Area

Formula

Distribution

Model

Seismic Intensity

Distribution**

Predominant

Types*

Spatial Distribution Small Program

for Personal

Computers

* Assigned from a nine-category classification (see the following section). ** Intensity distribution is given based on earthquake magnitude, epicentral distance, and site amplification.

Table 2: Earthquake disasters of which data were used in the verification of the model

Affected Country Year M Fatalities

1 Iran 1962 7+1/4 12,000 - 15,000

2 Turkey 1966 6.5 2,394

3 Turkey 1967 7.5 89

4 Iran 1968 7.3 7,000 - 10,000

5 Turkey 1970 7.1 1,086

6 Turkey I97I 6.0 57

7 Turkey 1971 6.7 878

8 Iran 1972 7.1 5,000

9 Nicaragua 1972 6.6 5,000 - 11,000+

10 Turkey 1975 6.7 2,385

1 1 Guatemala 1976 7.5 22,778

12 China 1976 7.8 242,419

13 Turkey 1976 7.4 3,840

14 Algeria 1980 7.3 2,263

15 Italy 1980 6.8 2,735 - 4,689

16 El Salvador 1986 5.4 1,500

seismic regions, Stadia Geoph. et Geod.,

Vol. 9, pp. 236-249

Shiono, K., F. Krimgold, and Y. Ohta (1991): Post-event rapid estimation of earthquake

fatalities for the management of rescue

activity, Comprehensive Urban Studies,

Center for Urban Studies, Tokyo

Metropolitan University, No. 44, pp. 61-106

382
Table 3: Ratio of fatality rate in various construction types to that in brick buildings

Building Type Ratio of Fatality Rate*

Field stone, Adobe 2.2 Brick, Cut stone 1.0 Poor-quality reinforced concrete 0.33

Wood frame with thick infill walls 0.12

Wood frame with thin infill walls 0.026

Wood 0.0017

* Fatality rate in brick building is used as the standard.

Table 4: Assumed built and demographic environment of the Kocaeli area

Construction Type Relative Number of

Buildings

Relative Number of

Occupants*

Population

Density

Brick 40% 10.1% 24.5

Half-Timber 30% 8.1% 18.4 Poor-Quality RC 30% 81.8% 184.1

Total 100% 100% 227**

* Ten (10) was assumed as the number of families residing in a reinforced concrete apartment

building. ** The population density of the affected area was given at 227 persons per sq-km, which is the population density of Kocaeli Province.

383
Construction Type

\

Vulnerabiltiy

Function

DS=f(l, CT)

\

Collaps

Fun(

CR=f(

.e Rate

::tion

1, CT)

Spatial Distribution of Fatality Rate

Magnitude

Max (Epicentrai)

Seismic Intensity

lo=f(M)

\

Attenuation Function

l=f(lo, r)

"S

Site Effect Population Density

Seismic Intensity Distribution (Circular

Isoseismals)

Affected Area

Fatality Rate

Function

FR=f(CR, CT) Spatial Distribution of Population

Number of Deaths

r: Epicentrai Distance

I: Seismic Intensity

lo: Maximum (Epicentrai) Intensity M: Magnitude

CT: Construction Type

DS: Damage Score CR: Collapse Rate

FR: Fatality Rate

Input Data and

Result (Estimate)

Knowledge

(Relationship)

Transmitting

Indicators

Figure 1 : Flow of fatality estimation in the distribution model.

384
D = ^ i (p V (\)) d r d e

D ; Total deaths

f : Deaths /sq-km

I ; Seismic intensity V : Collapse rate of buildings p ; Population density

Isoseismals

Figure 2: Schematic presentation of fatality estimation in the distribution model.

Collapse Rate (%

Field Stone, Adobe till Brick, Cut Stone Wood Frame

(Thin Infill Wall) Wood Frame

(Thick jfl^ijl Wall) Wood

(Wood Wall) Poor-Qi^lity RC

Good-Qualitv RC

7 8 9 10 11 Seismic Intensity (MSK)

Figure 3: Collapse rate functions.

Fatality Rate (%)

All Construction Types

Except the Two Below

Wood Frame

(Thin Infill Wall)

Wood Frame

(Wood.Wall)

0 20 40 60 80 100 Collapse Rate (%)

Figure 4: Fatality rate functions.

385
Estimated Deaths

1,000.000

100,000

10,000 r

1,000

100

10

R=0.84 f

R=0.90 (D>1,000) /

Reported Deaths

Figure 5: Comparison between reported and estimated death tolls.

Deaths

1,000,000

5 6 7 Magnitude

Y=9.0E-2 exp(1 .50X) (R*R=0.22)

Figure 6: Relationship between earthquake magnitude and fatalities; Crude death tolls are plotted.

386
Deaths

1,000,000

Magnitude

Figure 7: Relationship between earthquake magnitude and fatalities; Death tolls are normalized by the global population density.

Deaths

100,000

Field Stone, Adobe

Brick, Cut Stone

Poor-Quality RC

Wood Frame

(Thick Infill Wall)

Wood Frame

(Thin Infill Wall)

Wood Frame

(Wood Wall)

Population density was

given at 100/sq-km

Figure 8: Relationship between earthquake magnitude and fatalities; Population density is normalized at 1 00 persons per sq-km.

387
Deaths

1,000,000

Magnitude

Figure 9: Relationship between earthquake magnitude and fatalities; Death tolls were normalized by population density and corrected by building type.

Estimated Deaths

1 ,000,000

100.000

10,000

1,000

100

10

X•

* K

$

ofi o • t / «

• /

J •

1 lllllll 1 . . . . ....1

Distribution Model

(R=9.84)

Point Model

(R=9.73)

Reported Deaths

Figure 10: Comparison between recorded and estimated death tolls; Death tolls estimated in the two models can be compared.

388
Reinforced Concrete - 30%

(Poor Quality)

Brick - 40%

Wood Frame with Thick Infill Wall - 30%

(Half-Timber)

Figure 1 1 : Relative number of buildings by construction type in the affected area of the Kocaeli earthquake.

Deaths in an M7.4 Earthquake

30,0001

Figure 12: Increase in expected deaths with the progress of regional development.

389

Predictors For People's Response to a Tornado Warning, Arkansas, March 1, 1997

By

Lina Balluz, MPH, Sc.D.S Laura Schieve, MPH, Ph. D

. ^, TaImage Holmes,

Ph.D.\ Stephanie Kiezak, MS^ /Josephine Malilay, Ph.D.^

ABSTRACT

Nine powerful (F3-F4) tornadoes hit Arkansas

from the southeastern to the northeastern part

of the state during the afternoon of March

1,1997. The strongest of the tornadoes touched

down in Saline and Clark Counties. We

assessed factors associated with people's

response to the tornado warning issued in Clark

and Saline Counties, Arkansas. We conducted

a population-based cross-sectional study in Clark and Saline Counties. We administered a comprehensive questionnaire by telephone to

all selected individuals who stated they were at home prior to tornado touch down. Factors associated with response to the warning were

educational level equivalent to or greater than

high school , presence of a basement in the

house , hearing a siren , and having prepared a plan to respond when tornadoes occur. A

summary of the results and recommendations

are presented.

KEYWORDS : Cross-sectional study. Disaster

response. Emergency management, Tornado,

INTRODUCTION

Approximately nine powerful tornadoes hit Arkansas from the southwest to the northeast

part of the state during the afternoon of March

1, 1997 (National Weather Service, 1997).

Damage indicated that two counties (Saline and

Clark) had been hit by tornadoes of F4 intensity on the Fujita-Pearson tornado scale (National

Weather Service, 1997). These tornadoes were

estimated to be one half to one mile wide, with

wind speeds in excess of 200 miles per hour

(National Weather Service, 1997). Twenty-six

The most common sources from which

participants learned about tomadic activity are

fatalities and 400 nonfatal injuries were reported

(CDC, 1997). Tornadoes are the most frequent

adverse weather event likely to result in a

disaster in the United States (Noji, 1997). These

storms can result in mass-casualty situations for emergency and local medical facilities (Harris,

1992; Glass, 1980; CDC, 1985). Inadequate

warning for the population at risk is one of the

main contributing factors to tornado-related

deaths and injuries. However, little information

is available about the public's response to these

warnings, such as the percentage of people

seeking appropriate shelters after hearing a tornado warnings.

In this study, we present the results of a posttornado survey of community residents in Arkansas to determine factors associated with

appropriate responses to tornado warnings.

METHOD AND RESULTS

We conducted a population-based crosssectional study of residents of Clark and Saline

Counties. These counties were chosen because

they sustained the most severe property damage,

and experienced the most fatalities and injuries as a result of the tornadoes. We used randomdigit dialing to select a sample of residents from

these counties. We administered a comprehensive questionnaire to the 146

participants. Of the 146 participants, none were

injured during the tornado, and 5 (3.4%)

reported damages to their homes. In addition,

140 (95.9%) understood the terms "tornado

watch" and "tornado warning". Of those 140

participants, 64 (45.7%) responded to the

warning by seeking shelter.

listed in Table 1 . People with at least a high

school education were more likely to respond to

391
warning (OR=4.2, 95% CI= 1.1-15.5). People

living in a house with a basement were more

likely to respond than those without a basement

(OR=3.8, exact 95% CI=1.1-17.1). In addition,

people who responded were more likely to live in an area actually hit by a tornado (0R= 7.2,

95% CI= 1 .9-40.6), to have heard sirens

(OR=4.4, CI=1.3-18.9) and to have prepared a plan of action (OR=2.6, 95% CI=l.l-6.3) than were those who did not respond to the warning.

We also looked at when participants first learned of tornado activity and the amount of

time they had to respond. Of the 146

participants, 82 (56%) reported first being aware of a tornado threat at least 10 minutes

before it passed through the area, and of the 64

who responded to the warnings, 58 (90.6%)

reported a response time (time between hearing a warning and seeking shelter) of 5 minutes.

CONCLUSIONS:

Our findings are consistent with other postdisaster investigations that identified factors

specifically associated with reducing mortality

and injury. We showed that those factors were

also associated with seeking appropriate

shelters after hearing tornado warnings.

We recommend that people who live in tornado-prone areas without underground

shelters, should prepare a plan of action to help

them respond immediately to tornado warnings

in an effective manner. Elements of this plan

may include engaging in tornado drills to familiarize members of the household with

safety procedures, sources of shelters in the

home or outdoors, and keeping specialized

weather radios such as those from the National

Oceanic and Atmospheric Administration

(NOAA) available in the home. Because the

frequency of tornado watches may desensitize

people to the dangers of tornadoes, we also

recommend increased public health education

about how to respond to tornado watches and

warnings in areas with high tomadic activity.

When planning protective actions such as constructing accessible tornado-resistant

shelters, and developing tornado-preparedness

programs for at-risk populations, emergency

management officials should keep in mind that

people have limited time to find shelters. References

1 . Centers for Disease Control and Prevention

(1997) Tornado-associated fatalities-Arkansas,

1997. MMWR 46(no.l9).

2. Centers for Disease Control and Prevention.

(1985) Tornado disaster. North Carolina, South

Carolina. MMWR 34:205-212.

3. Glass RI, Carven RB, Bergman DJ, et al. (1980) Injuries from the Wichita Falls tornado:

implications for prevention. Science 207:734-

738.

4. Harris LF. (1992) Hospitalized tornado

victims. Ala Med 61:12-16.

5. Noji EK. (1997) The public health

consequences of disaster. Oxford University

Press.

^Centers for Disease Control and

Prevention, National Center for Environmental

health, 1600 Clifton Road, NE, MS E-23,

Atlanta, GA, 30333

392
Table 1 . Sources of tornado warnings heard by survey

participants who did and did not seek shelter Arkansas, 1997

Sources Sought shelter Did not seek shelter

Number Percent Number Percent

Television 52 81.2 59 78.6

Siren 46 71.8 42 56.0

Weather changes 32 50.0 30 40.0

Commercial radio 16 25.0 8 10.6

393

POST-DISASTER CARBON MONOXIDE POISONING: AN EMERGING HEALTH PROBLEM

by

W. Randolph Daley, DVM, MPRI

ABSTRACT

Carbon monoxide (CO) is an odorless, colorless

gas capable of causing acute injury and death.

Following meteorological disasters over the past

decade, CO poisoning has become an

increasingly recognized problem. Through case

descriptions and studies of post-disaster CO

poisoning, important risk factors can be

recognized and controlled.

Major sources of CO poisoning vary greatly,

largely dependent on the type of the disaster. A

number of CO sources were important in poisoning cases following recent winter storms

in Washington, Tennessee, New York, and

Maine. These included gasoline-powered

electric generators, charcoal grills, portable gas

stoves, and portable space heaters. Following snow storms in Massachusetts, New York, and

Pennsylvania, cases of CO poisoning resulted

from exposures in idling motor vehicles.

Several severe poisoning cases in North

Carolina after Hurricane Fran and the first two

deaths related to Hurricane Georges in Puerto

Rico were due to CO poisoning from generators.

Investigations of these outbreaks of CO

poisoning have greatly increased our

understanding of the important factors involved.

By properly applying this knowledge, future

morbidity and mortality from post-disaster CO

poisoning can be prevented.

KEYWORDS: carbon monoxide, poisoning,

disaster, winter storm, hurricane, epidemiology,

environmental exposure.

^U.S. Centers for Disease Control and Prevention, National Center for Environmental Health;

Mailstop E-23, 1600 Clifton Road, Atlanta, GA 30333

395
1.0 INTRODUCTION

CO is a byproduct of the incomplete combustion

of carbon-based fuels. On inhalation, it causes poisoning through a number of mechanisms, but

chiefly by interfering with the oxygen-carrying

capacity of hemoglobin in the blood. Treatment

consists of providing supplemental oxygen. In severe cases this oxygen should by supplied

under hj^jerbaric conditions.

CO is responsible for more fatalities in the

United States each year than any other toxicant.

During 1979-1988, from 878 to 1513 deaths per

year were attributed to unintentional CO

poisoning. Additionally, from 1992 through

1996, an estimated 9,800 cases of CO poisoning

One of the first mentions of a CO poisoning

outbreak following a weather-related disaster

occurred after the great blizzard of February

1978 in eastern Massachusetts. Of 27 stormrelated deaths identified by mortality

surveillance, 5 were due to CO poisoning in stranded motorists. These individuals

presumably continued to run their automobile

engines to keep warm.

Nine years later, in October 1987, an outbreak

of CO poisoning was again documented

following a severe snow storm striking upstate

New York. In fact, CO poisoning was the most

common storm-related illness diagnosed from

the five emergency departments participating in

this investigation. It accounted for 22 of 59

non-traumatic storm-related cases seen during

the storm and the five following days. The

identified CO sources included electric

generators, kerosene heaters, and charcoal

briquettes used indoors to provide power or

heat. It is of note that in this outbreak the occurrence of a cluster of CO poisoning cases was unexpected by the investigators and

considered an unusual finding.

Idling motor vehicles again caused CO

poisoning after a blizzard in the northeastern

United States in January 1996. From January 8

involving consumer products were treated

annually in hospital emergency rooms. However, the true number of cases seen by

emergency physicians is probably far higher.

Though CO exposure is a year-round problem,

incidence of poisoning death increases during

the winter months. Most unintentional CO

poisoning occurs sporadically as a result of

motor vehicle exhaust or improperly functioning

home heating units. However, outbreaks have

occurred increasingly over the last decade

following winter storms and other disasters.

2.0 SNOW STORMS

to 9, in Philadelphia, Pennsylvania, CO

exposure caused two severe poisoning cases and one fatality. Over the same time period in New

York City, 21 severe CO poisoning cases and 1 CO death occurred. In every case the victim

became exposed while inside an idling motor

vehicle which had an exhaust pipe blocked by

snow. The victims were not reported to be

stranded but had started their cars before

clearing the snow from around them.

3.0 WINTER WIND AND ICE STORMS

During an ice storm in March 1991, 55 cases of

CO poisoning were diagnosed in an emergency

department in Rochester, New York. Ofthese,

16 patients became unconscious. All but one of

the patients were treated and released from the

emergency department; the other patient was

transferred for hyperbaric oxygen therapy. In 67% of cases generators were the source of the

CO and in another 17% the CO source was a

fiiel-powered heater or stove. Other sources

included charcoal grills, fire places, and house

fires. A major outbreak of CO poisoning in western

Washington State occurred after a severe winter

wind storm disrupted electric power service to over 700,000 residents in January 1993. Wind

396
gusts from this storm reached 88 miles per hour

and near freezing temperatures followed.

Investigators reviewed records from 53 hospital

emergency departments in western Washington.

During the day of the storm and the three

subsequent days, 8 1 laboratory-confirmed cases of CO poisoning were identified. These cases represented 30 exposure incidents with a range

of one to nine patients per incident. In 1 5 of

these incidents charcoal was the sole or contributing source of CO. Another 12

incidents involved a generator. All patients

poisoned by generators were non-Hispanic

Caucasians. Almost all patients poisoned by

charcoal were Asian or Hispanic, and most did

not speak English. No fatal cases were reported

The largest documented post-storm outbreak of

CO poisoning occurred in Maine after the severe

ice storm of January 1998. This storm disrupted

electric power service to over 600,000 residents, some for over two weeks. There were 100

laboratory-confirmed cases identified from four

emergency departments in the central region of

the state. Cases were documented for two

weeks following the storm's onset but most

occurred during the initial week of the disaster. One patient died and five received hyperbaric

oxygen. The cases represented 42 separate

exposure incidents with a range of one to eight cases per incident. A generator was the sole or

contributing CO source in 30 incidents involving

85 poisoning cases. Eight incidents involved

kerosene space heaters, four involved propane

space heaters, and two involved portable gas

stoves. Charcoal grills were the sole source of

CO in two incidents. Six incidents involved two

CO sources.

In this outbreak, it was noted that around a third

of the population in this region of Maine used a

generator during the disaster. Many of these

operated the generator inside a garage. Most

people who operated generators during this disaster obtained it after the ice storm and had no previous operating experience. Among

families using generators, generator location was the strongest risk factor for poisoning.

and 43% of cases were treated with hyperbaric

oxygen. Warnings issued in the broadcast and

print media may have limited the severity of this epidemic.

In Nashville, Tennessee, 13 cases of CO

poisoning were identified from an emergency

department following an ice storm in February

1994. None of these patients died and five

received hyperbaric therapy. Six patients were

poisoned by charcoal grills, four by generators,

and three by a fuel powered heater or stove. Over half of patients were Asian in a city with a very small Asian community. Most patients were not fluent in English.

Households operating a generator inside a garage or other structure attached to the home

were at 1 9 times greater risk of becoming

poisoned by CO. Households operating a generator inside a basement were at 300 times

greater risk.

4.0 HURRICANES

CO poisoning has also been noted as an adverse

health effect of tropical storms. During the four

days immediately after Hurricane Fran in September 1996, 13 cases of CO poisoning were

seen in a hyperbaric treatment center in North

Carolina. The storm caused loss of electric

power to large areas of the state. These cases represented five exposure incidents. All

incidents involved use of a generator operated

indoors or outdoors near a ventilation source for a home.

In September 1998, Hurricane Georges

disrupted electric power service to most of the

island of Puerto Rico. Two days following the

hurricane a fatal case of CO poisoning occurred

due to a generator operated inside a home. The

next day another CO fatality resulted from a generator operated outside but near an opening

to a store. CO poisoning was not considered a major risk in Puerto Rico at that time. These

deaths represented the first two cases of

397
hurricane-associated mortality following

Georges in Puerto Rico.

5.0 CONCLUSION

In the United States, CO poisoning has emerged

as a major adverse health effect following

disasters. Though especially common after winter storms, it is also becoming a posthurricane problem. The potential for outbreaks

of CO poisoning exists in any disaster resulting

in widespread loss of electric power.

Gasoline-powered electric generators have been

the CO source in many post-disaster poisoning

incidents. Operation of generators inside a home or attached structure is especially risky,

but cases also occur from generators operated

Cases of post-disaster CO poisoning are almost

always preventable. In the aftermath of severe winter storms or any disaster resulting in widespread loss of electric power, public service

announcements should be issued concerning the

dangers of CO poisoning. These should cover

the symptoms of CO poisonings and ways to prevent exposure. Any message should be

issued in language appropriate to local

communities. Because CO poisoning is easily

misdiagnosed, alerts should be sent to medical

providers and emergency response personnel

detailing diagnostic and therapeutic measures.

Interventions connected with CO sources such

as warnings, training, and CO detector

promotion should also be valuable. Unless

preventive measures are taken, outbreaks of

post-disaster CO poisoning may become

increasingly more common.

6.0 REFERENCES

Ault K. Estimates oi non-fire carbon monoxide

poisonings and injuries. Washington, D.C.: U.S.

Consumer Product Safety Commission 1997.

Centers for Disease Control and Prevention.

Carbon monoxide poisonings associated with

snow-obstructed vehicle exhaust systems -

outside. Increasing availability of generators to the general population in the United States may

play a major role in the emergence of postdisaster CO poisoning.

Other CO sources have been important

following winter storms. Indoor use of portable

ftiel-powered heaters and stoves has caused CO

poisoning after numerous winter disasters.

Charcoal is a potent producer of CO and is very

dangerous if used inside. Charcoal grills have

frequently caused CO poisoning following

winter disasters, especially among Asian and

Hispanic Americans. Exhaust from snowobstructed automobiles has been a particular

danger after snow storms.

Philadelphia and New York City, January 1996.

MMWR 1996;45:1-3.

Centers for Disease Control and Prevention.

Deaths associated with Hurricane Georges - Puerto Rico, September 1998. MMWR

1998;47:897-8.

Cobb N, Etzel R. Unintentional carbon

monoxide-related deaths in the United States,

1979 through 1988. JAMA 1991;266:659-63.

Cohen BA, Stolp BW, Dear BD, Moon RE,

Frazier LM. Carbon monoxide poisoning in the

aftermath of Hurricane Fran [letter]. Am J Public Health 1999;89:112.

Daley WR, Smith A, Paz-Argandona E, Malilay

J, McGeehin M. An outbreak of carbon

monoxide poisoning after a major ice storm in Maine. J Emerg Med 2000;18:87-93.

Geehr EC, Salluzzo R, Bosco S, Braaten J, Wahl

T, Wallenkampf V. Emergency health impact of

a severe storm. Am J Emerg Med 1989;7:598-

604.

Glass RI, O'Hare P, Conrad JL. Health

consequences of a snow disaster in

398
Massachusetts, February 6, 1978. Am J Public

Health 1979;69:1047-9.

Houck PM, Hampson NB. Epidemic carbon

monoxide poisoning following a winter storm. J EmergMed 1997;15:469-73.

Wrenn K, Connors GP. Carbon monoxide

poisoning during ice storms: a tale of two cities. J EmergMed 1997;15:465-7.

399

Assessment of Public Health Needs Among Displaced Persons Residing in a Temporary Camp

Following the August 1999 Earthquakes in Turkey

by

Dahna Batts-Osbome, MD, FACEPl

Lina Balluz, ScD, MPH

Mary Naughton, MD, MPH

ABSTRACT

The Centers for Disease Control and Prevention

(CDC) conducted a rapid epidemiologic

assessment of the needs of displaced persons

affected by the earthquake that struck western

Turkey on August 1 7, 1 999, and to evaluate the

morbidity that resulted from the disaster. The

results of two activities conducted in one

severely affected area to assess the public health

implications of the disaster for displaced persons

in two camps are presented.

KEYWORDS: Turkey earthquake, needs

assessment, public health, surveillance

1.0 INTRODUCTION

At approximately 3:02 a.m. on August 17, 1999, an earthquake measuring 7.4 on the Richter

Scale struck western Turkey and resulted in official estimates of 17,000 deaths and 10,000

people reported missing. Approximately 24,000

people were injured, and an estimated 600,000

people were left homeless. Numerous

aftershocks occurred during the following

month, causing further damage and loss. The

epicenter was approximately 7 miles southeast

of the city of Izmit, near the town of Golciik.

This location indicates that the earthquake

occurred on the northernmost strand of the

Morbidity data during the postdisaster phase at Bahcecik camp clinic and Izmir camp clinic in the Golciik area was characterized. Logbooks

North Anatolian fault system. The earthquake

originated at a depth of 10.5 miles and produced

at least 37 miles of surface rupture. A

community needs assessment was conducted in one camp in addition to a study of clinic visits in two camps, 2 and 6 weeks after the earthquake.

The purpose of the assessment was to provide an

objective postdisaster measure of needs to decision makers in the affected area.

2.0 METHODS

2. 1 Community Needs Assessment

A needs assessment was conducted in Bahececik

camp, which had been established by local authorities 1 week after the earthquake in the

Golciik region. Golciik was thought to be the area most affected by the disaster. This camp

had 248 tents. On October 1-2, a household

survey using a systematic random sample was

conducted, targeting a total of 1 55 households.

A household was defined as a unit of people

residing in one tent. One adult from each

selected household was interviewed using a standardized questionnaire that focused on

demographics, illnesses, injuries, sanitation,

shelter, and medical needs.

2.2 Morbidity Review

were reviewed for two 8-day periods: from

August 30 (the first day for which clinic records were available at both sites) through September

'Medical Epidemiologist, Centers for Disease Control and Prevention, Atlanta, Georgia

30333, dtb7(^cdc.gov

401
6, 1999, and from September 25 through

October 2, 1999.

3.0 RESULTS

3.1 Results of Community Needs Assessment

In Bahcecik camp, 154 households were visited,

and interviews were completed for 86 (56%).

The survey represented 339 persons with a median household size of four people. Of the 86

households, seven (8%) had a child aged 2 years or younger; nine (10%) reported a household

member aged 65 years or older; and three

households (4%) reported a pregnant female.

Fifty-four (63%) reported damaged and

uninhabitable structures, and 22 (26%) reported

homes that were completely destroyed.

Of the 86 households, 20 (23%) had a household

member who had sustained an injury, and 90%

of the injuries were lacerations. Sixty-nine

households (80%) reported having at least one

ill household member since the earthquake;

these 69 households reported a total of 128 ill people. Approximately 32 people (25%)

reported depression, but only 14 (44%) sought

treatment; 24 (19%) reported respiratory illness,

and 23 (96%) requested medical treatment. Twenty (17%) noted chronic diseases,

specifically kidney problems, hypertension, and

heart disease, and 16 (80%) sought treatment. Thirteen (10%)) noted gastrointestinal illness,

and 1 1 (86%) sought treatment. The Bahcecik

camp clinic provided medical care for people in 85 (99%) of the surveyed households. Of the 86

households, one (1%) reported an earthquakerelated death. This low percentage most likely

reflects the fact that Turkish families generally

live together in one household and that whole

The impact of a disaster on a population varies

with the type of disaster. An earthquake of great

magnitude is one of the most devastating events

in nature. The earthquakes in Turkey caused

extensive infrastructure damage and losses

worth approximately $6.5 billion. Rapid needs

assessment of the affected population is an important initial step of response (1). Results of

such an assessment can minimize inappropriate

families either died or survived. It does not

reflect the numerous friends, co-workers, and

extended family members who were killed.

Piped water was the major source of drinking

water, as reported by 8 1 (94%) households.

Seventy-five (87%) of the 86 Bahcecik camp

households reported that food was available and was mainly provided by the relief workers in the

camp. In 77 (90%) households, people had

access to transportation; 83 (97%) households

had garbage disposal by municipal collection;

and 83 (97%) households had access to showers.

Field latrines connected to septic tanks were most commonly used for human waste disposal

by households (83 [97%]) and were reported as "clean" or "somewhat clean" in 45 (52%.)

households. Electricity was not available for 79

(92%) households.

3.2 Results of Morbidity Data

There were 468 logbook entries at Bahcecik

camp clinic and 534 entries at Izmir camp clinic

for the 8-day period from August 30 through

September 6, 1999. During the second 8-day

period, September 25 through October 2, 1999,

41 1 and 669 entries, respectively, were reviewed

(Table 1). The overwhelming number of visits were for illnesses, rather than injuries.

The primary illness for both time periods in both

tent cities was upper respiratory tract illness,

followed by musculoskeletal pain. All other

causes, including diarrhea, represented no more

than 10% of the total clinic visits (Tables 2 and

3).

4.0 CONCLUSIONS

relief in terms of delays and content. In Bahcecik camp, where 88% of the population was homeless, the primary need was housing.

Most inhabitants planned to remain in their tents

until they received houses. In addition, the low

frequency of gastrointestinal diseases implies

that sanitary conditions at the camp were well

maintained. This information may control the rumors of epidemics and help local health

402
authorities focus on allocating resources to reduce crowding, to provide counseling, and to supply appropriate medications, which were

public health needs as reflected by the

predominance of respiratory diseases and

reports of depression. Although 73 (85%)

households indicated access to a source of

medication, and direct observation showed a well-stocked medication supply area, the most

common medications that were needed for diabetes, hypertension, and depression, as well

as analgesics and vitamins, were not available,

according to persons interviewed.

Other recommendations to the local health

authorities based on these activities included

providing shelter, heat, and clothing suitable for winter conditions; providing mental heahh

services, social activities, and community jobs to address the increased level of stress in the

communities; continuing the current level of

medical care; and encouraging reporting of

morbidity data from local camp clinics to regional health offices.

5.0 REFERENCES

Guha-Sapir D. Rapid assessment of health

needs in mass emergencies: review of

current concepts and method. World Health

Stat Q 1991;44:171-81

403
TABLE 1 MORBIDITY BREAKDOWN AT BAHCECIK AND IZMIR CAMPS, TURKEY, 1999

Tent City Illnesses (%) Injuries (%) Total

Bahcecik

Aug.30-Sept. 6, 1999 434 92.7% 34 7.3% 468

Sept. 25-Oct. 2, 1999 382 92.9% 29 7.1% 411

Izmir

Aug.30-Sept. 6, 1999 492 92.1% 42 7.9% 534

Sept. 25-Oct .2, 1999 628 93.9% 41 6.1% 669

TABLE 2

MORBIDITY AT BAHCECIK CLINIC 2 AND 6 WEEKS AFTER EARTHQUAKE, TURKEY 1999

DIAGNOSIS BAHCECIK WEEK 2 DIAGNOSIS BAHCECIK WEEK 6

Upper respiratory tract infection116(24.8%) Upper respiratory tract infection

125 (24.8%)

Musculoskeletal pain 32 (6.8%) Musculoskeletal pain 25 (6.0%)

Watery diarrhea 28 (6.0%) Skin Infection 17 (4.1%)

Psychiatric Illness 27 (5.8%) Dental/oral disease 16 (3.9%)

Hypertension 21 (4.5%) Hypertension 14 (3.3%)

TABLE 3

MORBIDITY AT IZMIR CLINIC 2 AND 6 WEEKS AFTER EARTHQUAKE, TURKEY 1999

DIAGNOSIS IZMIR WEEK 2 DIAGNOSIS IZMIR WEEK 6

Upper respiratory tract infection

126 (20.0%) Upper respiratory tract infection168 (21.8%)

Musculoskeletal 61 (9.6%) Musculoskeletal 52 (6.8%)

Skin rash 41 (6.5%) Skin infection 24 (3.1%)

Hypertension 40 (6.3%) Dyspepsia 21 (2.7%)

Lower respiratory tract infection35 (5.5%) Lower respiratory tract infection21 (2.7%)

404
Wind Storm-related Disaster Deaths in the United States, 1994-8

by

Julie Jacobson. MD

ABSTRACT

Disasters are responsible for deaths every year

in the United States. The Center for Disease

Control and Prevention (CDC) has partnered

with the American Red Cross (ARC) to systematically collect comprehensive

information on disaster-related deaths to determine who is at risk in disasters. From

January 1, 1994 through December 31, 1998,

wind storms were responsible for a majority of

these deaths. We found that decedents from

wind storm disasters were more likely to be

males or the elderly; women and children were

not at increased risk. People occupying mobile

homes were at a higher risk than individuals

occupying single family homes, especially in tornadoes. Decedents in hurricanes were more

likely to die from drowning than individuals

killed in other wind storm events. By

characterizing decedents in wind storm disasters

we can work to prevent the unnecessary loss of

life from these disasters and focus prevention

and mitigation efforts at those at risk.

KEYWORDS: DISASTERS; MORTALITY;

HURRICANE; TORNADO

1.0 INTRODUCTION

Since 1986, the American Red Cross (ARC) and

the Center for Disease Control and Prevention

(CDC) have partnered to provide timely and

comprehensive information on disaster-related

deaths for public health use. As part of the

Federal Response Plan, the ARC is mandated by

Congress to provide mass care to all

Disasters cause thousands of deaths each year

around the world. The worldwide population at risk from disasters is increasing with more people

living in areas vulnerable to disaster events. (World Disaster Report, 1999) This phenomenon

is also observed in the United States. By 2010,

the number of people will double living in hurricane-prone and seismically active areas. (National Mitigation Strategy, 1995) As

advances in technology increase the accuracy of

weather prediction, predictions need to be

communicated to those at risk to decrease

morbidity and mortality. (Noji, 1997) By looking

at deaths that result from wind storms we can

learn who is at risk of death from these events and improve our ability to prevent adverse health

outcomes.

Between 1994-98 in the United States, a total of

861 disaster-related deaths occurred. Of these

deaths 92% (791/860) were related to a climatic

event; the largest number of deaths, 62%

(490/79 1 ), were caused by wind storm events. (Jacobson) This report provides a descriptive

summary and associated risk factors for death in wind storm disasters from January 1, 1994

through December 3 1, 1998.

2.0 METHODS

Presidentially-declared natural disasters. Through an agreement with CDC, the ARC

response contains a surveillance component, The

American Red Cross- Centers for Disease

Control and Prevention Health Impact

Surveillance System for Disasters, which collects information on morbidity and mortality resulting

405
from these disaster events. (Patrick, 1992).

ARC staff actively and passively collect data

through medical examiner and coroner's offices,

hospital surveillance, home visits, funeral home

information, news papers and word of mouth.

They then confirm the death through the medical

examiner, coroner, or funeral home and fill out a standard surveillance form that records

demographic information, the location and cause of death, how the death was related to the

disaster, and the death circumstances. As ARC

contacts the families of decedents to provide

services, additional information may be obtained

about the death from surviving family members.

Completed forms are forwarded to CDC, where

data are cleaned and entered into Epi Info 6.1 and Microsoft Excel for analysis. We described

deaths by gender, age, cause of death, and

disaster type. Data were compared to US 1 996

(median year) census data to calculate risk.

3.0 RESULTS

3.1 TYPE OF DISASTER

From January 1, 1994 to December 31, 1998, 65

wind storm-related deaths occurred in 4

categories; 23 involving tornadoes, 26

hurricanes (involving 7 tropical storms and 10

hurricanes), 12 flood/tornado mixed events, and

4 other storms. The largest number of deaths were attributed to tornadoes with 208 deaths,

followed by hurricanes with 194 deaths,

flood/tornado mixed events with 79 deaths, and

other storms with 9 deaths. Hurricanes had the

largest number of deaths per disaster with 14

deaths/disaster (tropical storms 8 deaths/disaster). Tornadoes had the second

most deaths with 9 deaths/disaster. The

deadliest disaster type however was tornado,

with a death rate of 7 per thousand people

We then compared the demographics and

location of death of decedents in the two most

severe wind disaster types, tornadoes and

hurricanes. (Table 2) Males had a higher risk of

affected, followed by other storms (2/1000

affected), flood/tornados ( 1/1000 affected), and

finally hurricanes (0.6/1000 affected).

3.2 DEMOGRAPHIC INFORMATION

During this 5-year period, 490 people died in wind storm disasters. Decedents had a median

age of 41 years (range <1 to 107 years) and were 62% male (299/486). Males had a relative risk of death of 1.73 (95% CI 1.44- 2.08) compared to females. Six percent (30/483) of decedents were

<5 years of age, 19% (92/483) were <1 8 years of

age, and 20% (96/483) were >65 years of age. Individuals aged >65 years had a relative risk of

death from disasters of 1.55 (95% CI 1.25- 1.93)

when compared to all disaster-related deaths. No

other age group showed a statistically significant

difference in death rate.

3.3 LOCATION OF DEATH OR FATAL

FNIURY (TABLE 1)

Deaths from wind storms occurred primarily in single family homes (25%)) and outdoors (24%i).

The remaining deaths occurred in mobile homes

(19%)), motor vehicles (19%)), multi-family

dwellings (<1%)), and other (including public

buildings, boats, etc). To calculate the relative

risk of location of death we compared mobile

homes and multi-family dwellings to single

family homes. The risk of death in a mobile

home was 7.89 times (95% CI 5.79- 10.76) that of being in a single family home. Occupying

multi-family homes actually had a reduced risk of

death 0.03 (95% CI 0.00- 0.18).

3.4 TORNADOES VS. HURRICANES

death in hurricanes (RR 2.47, 95% CI 1.81-3.37)

and a marginal increase in risk from tornadoes

(RR 1.32, 95% CI 1.00-1.74). Elderly (>65

years)in tornadoes, were the only age group with

406
an elevated risk of death (RR 1 .59 95% CI 1.13- 2.24). The risk of death in a tornado when in a mobile home was 1 1.61 times (95% CI 7.82-

1 7.23) that of being in a single family home.

This was not the case for hurricanes; no

statistical difference was observed between

occupying mobile homes and single family

homes. Compared to all wind storm-related

deaths, tornadoes had a higher risk of a death

from trauma (RR 1.28, 95% CI 1.1 1-2.15) and

hurricanes had a higher risk of drowning (RR

1.34, 95% CI 1.06-1.70).

4.0 DISCUSSION/CONCLUSIONS

Most disaster-related deaths in the United States are related to climatic events. Improvements in detection and warning systems over the past 4

decades have been credited with a steady

decrease in disaster-related deaths (National

Research Council 1994). The effects of these

warning systems may be reflected in the

differential in death rates by disaster type.

Disasters with the least amount of warning time,

tornadoes, have the highest death rate (7.2

deaths/ 1 000 affected) and disasters with more

warning time, hurricanes, have a lower death

rate (0.6 deaths/1000 affected) (Figure 2).

Tornadoes are the most lethal disaster type in the United States with 7.2 deaths per 1000

people affected. In 1952, the National Weather

service began broadcasting tornado warnings

and since that time the number of deaths from

tornadoes has been decreasing (Noji 1997).

However, over this 5 year period, the number of

deaths from tornadoes and flood/tornado events has risen upward (Figure 1). Historically, North

America has the most severe tornado disasters on the earth. (Bohonos 1999) Studies have

shown that warning time and access to proper

sheltering have the greatest impact on morbidity

and mortality (Bohonos 1999, Belville 1987,

Carter 1989, Eidson 1990). Occupying mobile

homes have repeatedly been shown to be a risk

for death. Our data support this with an increased

risk of death of 1 1 .61 times when compared to other residences. Therefore, prevention strategies

should be directed to 1) occupants of mobile

homes and 2) use of timely effective warnings.

Mobile homes in tornado prone areas should have

safety shelters or safe rooms. All households

residing in tornado prone areas should have a National Oceanic and Atmospheric Association

(NOAA) Weather Radio to warn them of storm

warnings. NOAA radios will turn themselves on when a warning is issued, so even at night or when families are asleep they will be alerted of

the risk of oncoming severe weather and be able

to take the proper precautions (NOAA 1999).

NOAA radios should be as common as smoke

detectors.

Hurricanes affected the largest number of people

(3 1 0,969) and had the third largest number of

deaths by disaster type (194) yet were the least deadly (0.6/1000 affected). However, advanced

warning systems should allow us to decrease this rate further. One of the most interesting findings

with hurricanes was the increased risk of

drowning when compared to other wind storm

disasters. Hurricanes are mixed events by their

very nature involving not only wind, but storm

surge and potentially heavy rains. The

Saffir/Simpson hurricane scale that is used to determine what areas are evacuated, is based

primarily on wind speed and storm surge. Given

the increased risk of drowning, some of these

deaths could potentially be avoided if rainfall and

flooding risk were incorporated into the rating

scale.

Males and the elderly are at higher risk of death

from wind storm disasters (RR 1 .73) than females

and all age groups, respectively. This refutes previous findings that women and children are at the greatest risk of death from disasters (Noji

1997). As in previous studies, our findings

confirm the increased risk of death for the elderly

in wind storm disasters as a whole (RR 1.55) and

407
in tornadoes specifically (RR 1 .59). However,

we did not find an increased risk for women and

children. Males were at an increased risk in all The ARC CDC Health Impact Surveillance

System for Disasters provides us with the first surveillance system for disaster-related deaths

that systematically collects data on all disasters

with detailed information on the decedent, the cause and circumstances of death in a single

data base (Patrick 1992, Storm Prediction Center

(SPC) 2000). Although this data may not

represent all disaster-related deaths, it is one of

the most comprehensive disaster data collection

systems. Although it may under-represent the

total number of deaths from small disasters and

localized events, this system it provides much

information on deaths resulting from large scale

disasters represented by all Presidentially

declared disasters and disasters with a national

ARC response.

By looking at deaths that result from disasters,

we can learn who is at risk of death from these events and improve our ability to prevent

adverse health outcomes (Binder 1987). As the

population at risk of injury and death from

disasters increases, we need to work to mitigate

and prevent these risks. By providing

communities at risk with the tools to protect

themselves including: 1) NOAA radios for warnings of severe weather and alerts 24 hours a day; 2) knowledge of flood risk areas and safe

evacuation routes prior to and during flood

events; and 3) safe rooms for people residing in mobile homes, we can decrease the unnecessary

loss of life in disasters.

6.0 REFERENCES

Binder S, Sanderson LM. The roll of the

epidemiologist in natural disasters. Ann Emerg

Med. 1987 Sep; 16(9): 1081-4.

Carter AO, Millson ME, Allen DE.

wind storm disaster types.

Epidemiologic study of deaths and injuries due to tornadoes. Am J Epidemiol. 1989 Dec;

130(6):I209-18.

CDC. Tornado-Associated Fatalities - Arkansas,

1997. MMWR 1997; 46(I9):4I2-416.

CDC. Risk factors for death in the 27 March

1994 Georgia and Alabama tornadoes. MMWR.

l995;l9(2):l70-7.

Federal Emergency Management Agency

(FEMA). National Mitigation Strategy

Partnership for Building Safer Communities.

FEMA. December 6, 1995

International Federation of Red Cross and Red

Crescent Societies (IFRC). World Disasters

Report. IFRC. 1999.

Morrow BH. Women and families in disaster: Emergency management issues and

recommendations. Proceedings of the Annual

International Conference of the Business

Administration, London. 1995 June.

Mulilis JP. Gender and earthquake preparedness

A research study of gender issues in disaster

management: differences in earthquake

preparedness due to traditional stereotyping or

cognitive appraisal of threat?. Austr J ofEmerg

Managment. 1999 Autumn; 14(1): 41-50.

NOAA. NOAA Weather Radio the voice of the

National Weather Service.

Http://www. noaa.gov/nwr/nwr.pdf.

Noji EK. The Public Health Consequences of

Disasters. Oxford University Press. 1997.

Belville JD. The National Weather Service

warning system. Ann Emerg Med. 1987

Sep;16(9): 1078-80.

408
Patrick P. The Red Cross and CDC's NaturalDisaster Surveillance System. American Journal

ofPublic Health. December 1992, Vol 82, No.

12.

Pielke RA Jr, Pielke RA Sr. Hurricanes their Nature and Impacts on Society. John Wiley &

Sons. 1997.

Sapir DG. Natural and man-made disasters: the vulnerability of women-headed households and

children without families. World Health Stat Q.

1993;46(4):227-33.

Storm Prediction Center (SPC). Tornadoes and

Deaths by Year and Month 1950-1997.

http://www.spc.noaa.gov/archive/tornadoes/ustd

bmy.html.
Table 1. Location of death or fatal injury in wind-storm fatalities in the United States,

1994-98

VV CdLllCl 1 CidLCU. tJCdLlio

Outdoors 24%

Single-family home 25%

Multi-family home <1%

Mobile home 19%

Motor Vehicle 19%

* remaining deaths occurred in the 'other' category

410
Table 2. Characteristics of decedents and location of death in tornadoes and hurricanes in

the United States, 1994-98

Demographic data Tornado Hurricane

Gender- male 56% 62%

Mean age 40 years 42 years

Location of death* n=396

Outdoors 18% 29%

Single-family home 29% 27%

Multi-family home 0% 0.1%

Mobile home 32% 5%

Motor Vehicle 7% 23%

Primary causes of death n=488

Trauma 85% 30%

-head trauma 17% 7%

Drowning 5% 42%

* remaining deaths occurred in the 'other' category

411

Theme 6

Natural Disasters

in the World

and the

413

URBANIZATION AND URBAN DISASTER PREVENTION

by

Tomomitsu FUJIl'^

ABSTRACT

In Japan, urbanization has been accompanied by

the appearance of increasingly advanced and

complex social systems. And disasters caused by

earthquakes, intensive rainfall and other natural

forces that caused property damage and human

casualties have evolved into quite complex urban

disasters that bring secondary disasters in addition

to the primary damage.

This report reexamines past urban earthquake and

flood disasters to study changes and features in urban disasters accompanied by urbanization. The

report also presents an outline of urban disaster prevention measures that must be promoted to prevent urban disasters and minimize their effects

to increase the safety of urban areas.

KEY WORDS: Urbanization

Urban Disaster

Earthquake, Flood

Disaster Prevention Measure

I. INTRODUCTION

The rapid industrialization and urbanization since

the beginning of high-speed economic growth in 1955 has concentrated the population in the three huge metropolitan regions around Tokyo, Nagoya,

and Osaka. Urbanization has been accompanied

by the appearance of advanced social systems and

facilities, and by their increasing close

interrelationships.

Once social systems and facilities are destroyed or disrupted by natural disasters, not only does the system itself fail, ripple effects cause chain

reactions triggering new forms of disasters that

did not seen in the past. In this way, disasters that occur in urban area have been transformed and

evolved as a by-product of urbanization.

Focusing on natural disasters accompanied by

urbanization, this report reexamines past

earthquake and flood disasters to study the changes, features and lessons of urban disasters. The report also includes an outline of urban

disaster prevention measures against earthquakes

and intense rainfalls that must be promoted to prevent urban disasters and minimize their effects considering recent social trends including

internationalization, information technology

revolution, and aging of society.

2. URBAN DISASTER

2.1 Urbanization

Figure 1 shows changing numbers of human

casualties due to past natural disasters. Although

continuous implementations of disaster prevention

1) Director-general, Public Works Research

Institute, Ministry of Construction

1 Asahi, Tsukuba, Ibaraki 3050804 Japan

415
measures has been rewarded with a steady fall in

the number of casualties, continued concentration

of properties and population in urban regions

increased their susceptibility against natural

disasters. And the 1995 Hyogo-ken Nanbu

Earthquake caused the most serious disaster in the

recent history of Japan. ^ 7000 c5 6000

O. 5000 2 4000 J 3000 O 2000

inO 1000

MIA -

1 nil Hh

mm

n lli^ii'o'i^.s'.!^ ^

AT Sr 55' 59' 63' 67' 71' 75' 79- 83' 87' 91' 95' Fiscal Year

Fig. 1 Human Casualties due to Natural Disasters

In urban regions, new social systems and facilities have appeared and spread to support the

socio-economic activities (Figure 2). Many of

these social systems and facilities have never been

exposed to a large-scale disaster and have the

potential to cause unpredictable forms of damage

and complex chain-reactions of secondary

disasters.

Background

Rising Price of Land,

Concentration ot

Population/ Functions

Rapid/ Mass

Transportation System

Infonnation Oriented

Improvement of

Living Condition

Facilities and

Functions

Multistory Buildings Underground Arcade

Reclaimed Land

Chemica] Plant Highway System Subway System

Online Computer System Water Supply/ Sewage

Convenience Store

Years

30' 50' 70' 90'

Major Earthquakes Kanto EQ Niigata EQ

Hyogo-ken Nanbu EQ

Fig. 2 Appearance of New Social Systems

and Facilities

Short

Affected Duration

Long

Rural Urbanizing Urban

Fig. 3 Evolution and Classification of

Natural Disasters (schematic)

2.2 Recent urban disasters

(1) Classification of natural disasters

Figure 3 schematically presents that natural

disasters have evolved by focussing on their

"effected duration", "effected region" and

"complexity of influence". This report categorizes

natural disasters as a "rural disaster", an

"urbanizing disaster" and an "urban disaster".

1) Rural disaster: This type means a classical disaster developed in a region where urbanization has not initiated and

neither infrastructures nor disaster prevention

facilities have been provided. Most of the effects caused are primary damage such as property

damages and human casualties. Effected diuration and region are limited.

2) Urbanizing disaster: This type of disaster occurs in rapidly urbanizing

416
regions where population has been increasing but

infrastructures and disaster prevention facilities have not been sufficiently provided. Rapid

urbanization in developing countries has increased

the frequency of this type of disasters around the world. Although effects caused by the disaster are mainly primary damages, effected duration and

region are larger than rural disasters.

3) Urban disaster: This type of disaster occurs in highly urbanized

metropolises and regional core cities. Where the

infrastructures are not sufficiently safe, extremely

severe primary damage occur, and cause

disruption of urban functions as the secondary

damage. Because urban functions are highly

interdependent and complexes, it is difficult to

predict the relationship between the external

natural force and the form of damages. Influence

of the disaster spread over wide area, and disaster

restoration requires long period comparing with

previous types of disasters.

(2) Recent urban disasters in Japan

1) Niigata Earthquake (1964/6/16, M 7.5)

Photo. 1 Failure of Highway Bridge due to Ground Liquefaction

Earthquake disaster developed in the urbanizing

Niigata Prefecture caused approximately 470 of

casualties, 1,960 of totally destroyed buildings,

and failures of highway bridges. Ground

liquefaction that occurred over wide areas of soft

soil deposit was an unpredictable phenomenon

that have never experienced before. And it was

pointed out that a ground cave-in and a lateral ground flow induced by the ground liquefaction

are likely to enlarge human and property damages.

Because most of core cities are placed on soft ground soil deposit in Japan, this disaster experience enforced countermeasures to prevent

the ground liquefaction.

2) Miyagi-ken-oki Earthquake (1978/6/12, M

7.4)

Earthquake disaster developed in an urbanized

regional core city of Sendai in Miyagi Prefecture

caused appropriately 1,350 casualties and severe damages on lifeline facilities for electric power/

gas/ water supply and sewerage. Damage of

buildings in newly prepared residential districts and human casualties due to the collapse of block

walls were never experienced in previous disasters. Because damages of lifelines influenced daily

lives and socio-economic activities for several weeks after the earthquake, upgrading the seismic

safety of lifelines was pointed out to be important.

Photo. 2 Damage of Power Plant

417
3) Nihonkai-Chubu Earthquake (1983/5/26, M

7.7)

Earthquake disaster developed in the urbanized

regional core city of Noshiro in Akita Prefecture

caused approximately 270 casualties, severe damage on transportation systems including

highways and railroads, and lifeline facilities. The

fact that about 75% of the casualties were the

direct result of Tsunami triggered by the

earthquake served as a sharp reminder of the

importance of prompt evacuation activities and

provision of information to residents following an

earthquake. Transportation systems and lifelines were seriously disrupted by damages to embankments and buried pipelines. Because it

takes a long time to find out and repair damage of

underground facilities, it was pointed out that seismic strengthening of underground facilities is

important.

Photo. 3 Up-lift of Underground Gas

Storage Tank

4) Intensive RainfaU in Nagasaki Pref. (1982/7/23,

187 mm/ hour)

Flood disaster developed in the urbanized regional core city of Nagasaki caused slope failures and

sediment flows around residential districts, and

disruptions of highway network. Approximately

1,100 casualties and 600 totally collapsed

buildings were developed. Because the City of

Nagasaki is located in a peninsula surtounded by

the sea, disruptions of highway network

temporarily isolated the city and obstructed rescue

activities. Even after the flood waters had receded,

mountains of debris and abandoned automobiles

blocked the streets so that it took a week just to re-open the trunk roads. The disaster was a sharp

reminder of the importance to secure the highway

network redundancy, and to re-open trunk road

promptly after flood disasters.

Photo. 4 Closure of Trunk Road due to Rood Debris

5) Hyogo-ken Nanbu Earthquake (1995/1/17, M

7.2)

Hyogo-ken Nanbu Earthquake was one of the

largest earthquake disaster developed in highly

urbanized region in Japan. Extensive ground

motion that was not formerly anticipated caused

approximately 5,500 fatalities, 93,000 totally

collapsed buildings, catastrophic damage of

transportation systems and lifelines, and damage

of other industrial facilities. Because disruptions

418
of arterial highways caused serious and long

lasting influence on nationwide transportation

activities, importance to provide alternate arterial

transportation routes was pointed out. The

importance of wide-area improvements to urban

districts was also illustrated by the spread of

uncontrollable fires that swept through districts with high concentrations of wooden dwellings.

Damage to disaster prevention bases and

information systems that seriously hampered the urgent response after the earthquake also reminded the importance to prepare appropriate

crisis management programs in addition to

disaster prevention programs.

Photo. 5 Disruption of Arterial Viaduct

6) Intensive Rainfall in Fukuoka Pref. (1999/1/29,

122 mm/ hour)

Rood developed in Fukuoka city caused 2

fatalities, and slope failure developed in Hiroshima Pref. caused 31 fatalities and 153

collapsed buildings. Aroimd the Hakata Station

where the use of imderground space was advanced, an underground arcade and underground station were flooded. In the basement of a nearby office

building, people who failed to evacuate promptly were drowned. The flooding cut off transportation

routes and caused the failure of power system

equipment blacking out the Hakata Station area. This flood clearly indicated that the city was

susceptible to the influence of intensive rainfall

that exceeded the anticipated amount of rainfall. The flooding of underground space was never

experienced before, and reminds the need of flood

disaster prevention measures for underground

space.

Photo. 6 Flooding of Underground Station

23 Features and lessons learned

(1) Features of urban disaster

1) Simultaneity and complexity

In addition to the damage of human lives and

properties, complex disasters including city fire, slope failure, tsunami, flooding, lifeline damage,

transportation damage, rumors, deteriorate of

environment, and other disasters are developed

simultaneously. And the form of disasters varies

according to topographical conditions, ground

conditions, and other unique features of regions.

2) Increase of human casualties

Highly concentrated populations tend to cause an

extremely increase of human casualties. Extensive

rescue and medical treatment activities are

required. Also numbers of emergency facilities for medical treatment and evacuation are needed.

3) Influence to socio-economic activities

419
Socio-economic activities in urban regions are highly dependent on infrastructures including

lifeline facilities, transportation facilities, information related facilities, etc. Once

infrastructures lost their functions, serious

repercussion may be induced. In the 1995

Hyogo-ken Nanbu Earthquake, for example,

disruption of expressway not only cause delayed

evacuations, delayed restorations works and

traffic congestions in the damaged region but

indirectly affected nationwide socio-economic

activities. And the damage of port facilities at Kobe not only cause loss of functions but also

affected socio-economic status in neighbor

countries through import/ export activities.

4) Chain-reaction and expansion of disaster

As a consequence of the highly interdependence

of urban functions, primary damage may induce

chain-reaction and cause the expansion of

secondary damage. For example, urban functions

are extremely dependent on electricity, failure of

power station affects almost all the urban

functions and socio-economic activities as the

secondary damage.

5) Difficulty of disaster prediction

Due to the complex relationships between urban

functions, damage that does not seem to be related with external natural force may be developed.

And due to recently appeared new facilities and

functions, unexpected new forms of damage may

be developed. In the 1971 San Fernando

Earthquake, for example, explosion of gas

pipeline and falling-off of curtain walls in

high-rise buildings were developed. In the 1978

Miyagi-ken-oki Earthquake, more than half of the

fatalities were induced by the collapse of block

walls. In the 1995 Hyogo-ken Nanbu Earthquake,

malfunctions of fire hydrants prevented

firefighters from halting the spread of city fires, and the damage of gas stations help leaving many

vehicles abandoned on city streets.

(2) Lessons learned from urban disasters

1) Importance of disaster mitigation planning

Although frequency of large-scale urban disasters

is quite low, disaster potentials is extremely high.

For promoting countermeasures and securing

safety of urban regions against such kind of "low

frequency and high influence" disasters, it is quite

important to develop appropriate disaster

mitigation plans that include risk management

measures for damage prevention and crisis management measures for minimizing disaster

influences by urgent response works. Also

construction of infrastructures must balance the

pursuit of advanced functions with consideration

of their safety and roles during disasters.

2) Assessment of growing susceptibility

It is essential to carefully study the credible

external natural force, and evaluate susceptibility

of urban regions. Susceptibility must be

appropriately evaluated from the following

viewpoints;

a) Distribution and attribute of population:

As the population concentrate to the urban regions,

disaster potential increases. As the population

increases, residents with various attributes also

increase, i.e. elderly, handicapped, foreigners, and

infants. As the population increases, relationship

in community is likely decline and residents with

little experience of disasters are increase. Housing

problem enforce long distance commute, and

induce huge population gap between daytime and

nighttime. Depending on the time disaster occurs, human casualties due to panic and traffic

accidents may extensively increase.

420
b) New facilities/ functions and chain-reactions:

As new facilities/ functions increase, new form of

damage and propagation of disaster due to

chain-reaction are likely developed. It is therefore

essential to evaluate disaster susceptibilities by

understanding vulnerability of facilities/ functions

and their relationships.

c) Spread of residential districts: Accompanied with the spread of residential

districts, residents living in risky districts from

disasters are increasing. Districts with extremely

high concentration of wooden dwellings have risk from city fire. Districts along steep slopes have

risk from slope failure. Districts with low

elevation have risk from flooding and tsunami.

And districts formed in reclaimed land have risk from ground liquefaction.

d) Advanced use of urban space:

Accompanied with the lack of land, elevated

structures have been constructed for utilizing

urban space. Although these structures are

necessary in urban regions, vulnerability and risk

from disasters must be carefully studied. For

example, High-rise buildings have risk from fire, underground facilities including arcade and

stations have difficulty in emergency evacuation,

and elevated railways and viaducts have risk from

earthquakes.

e) Hazardous plants and materials:

As the urban functions concentrated, hazardous

plants and materials to sustain urban activities are

required. These plants and materials include

energy production plants, pharmaceutical and

chemical plants, gas storage tanks, petroleum

supply bases, gas stations, etc. These plants and

materials have risk from explosions and serious

environmental problems.

3. DISASTER PREVENTION MEASURES

3.1 Basic concept

The "Disaster Countermeasures Basic Act" was

enacted in Nov. 1961 in response to a large-scale

disaster caused by the 1959 Ise-wan Typhoon. The

Act established the framework for disaster prevention measures in order to protect human

lives and properties from external natural forces including earthquakes, tsunamis, windstorms,

intensive rainfalls, landslides, floods, high tides,

volcanic eruptions, heavy snows, etc. Figure 4

shows the relationship between disaster prevention plans prepared for national level,

Prefectural level, municipal level and citizens'

level.

National Level: prepare, execute, coordinate basic plan

I Prime Minister

i Central Dbaster Prevention Council

; Designated Administrative Organs

^Basic Plan"

^Operational Plan'

Prefectural Level: execute, coordinate operations

"Prefectural Plan" Governor

Prefectural Disaster Prev. Council

Designated Local Organs

Cities, Towns Level: execute operations on site

"Municipal Plan" Mayors

Municipal Disaster Prev. Coondl

Citizens Level: participate in operations Citizens Level

Fig. 4 "Basic Plans for Disaster Prevention"

in Japan

Here, the disaster prevention plan was revised

following the 1995 Hyogo-ken Nanbu Earthquake.

Although disaster countermeasures have been

implemented to protect human lives and

properties from anticipated natural external forces,

the concept of crisis management to minimize the

disaster influence was additionally introduced.

This is because we must recognize technical

difficulties to predict a maximum limit of external

421
natural forces, and technical/ financial difficulties

to completely protect facilities/ functions from

damages against catastrophic levels of external

natural forces.

3.2 Earthquake disasters

(1) Basic concept for disaster prevention

Although a mechanism of earthquake occurrence have been investigated, it is technically difficult to

predict the maximum scale, accurate location and

time of earthquakes. And it is financially/

technically unrealistic to attempt protecting

facilities/ functions from damage considering

catastrophic levels of external earthquake force. In

the "Basic Plan for Disaster Prevention" described

above, basic concept for preventing damage of

structures/ facilities is described as follows (Table

1):

Table 1 Basic Concept for Preventing Damage

Earthquake Motions

Ordinary Motion

once or twice during lifecycle

Extremely

StrongMotioa

very low probability

during iifecyde

8

&

Ordinary

Structures

Significant Loss

of Functions must be

Prevented

Hireatto

Human Lives Must bePrevented

Important

Structures

Significant Influence To Urgent Activities and Sodal Activities Mustbe

Prevented

1) External earthquake forces

Two levels of earthquake motions must be

considered in the seismic design of structure/

facility. The first level is ordinary motion that has

certain probability to strike structure/ facility once

or twice during its lifecycle. The second level is extremely strong motion that has very low

probability to strike structure/ facility during its

lifecycle. The second level of motion includes that generated by interplate earthquakes in the ocean, and that generated by earthquakes due to inland

faults.

2) Requirement of safety

Significant loss of functions must be prevented

against the first level of ordinary earthquake

motion. And serious threat to human lives must be

prevented against the second level of extremely

strong earthquake motion. In addition, sufficient seismic safety must be provided to structure/

facility that failure of function affect significant

influence to urgent response activities and

socio-economic activities. The amount of safety

must be determined by considering importance of

each structure/ facility.

For securing safety of structure/ facility, other

than the individual seismic design, it is also important to secure function of systems by

increasing system redundancy and providing

alternatives.

(2) Disaster mitigation measures

After the 1995 Hyogo-ken Nanbu Earthquake, the

Ministry of Construction compiled the "Concept

of Urban Planning emphasizing on Seismic

Safety". Based on the concept, disaster mitigation measures have been promoted. In addition to the

seismic design/ retrofitting of infrastructures/

buildings, following principal policies have been

strongly focused.

1) Reduction of vulnerability

The policy includes the reconstruction of

vulnerable districts with high concentration of

wooden dwellings, establishment of disaster

proofing zones, and distribution of open spaces to

prevent the spread of city fire.

422
2) Improvement of core public facilities For securing urban functions for evacuation, fire

fighting and emergency transportations, the policy

enforcing implementations of the evacuation route network, emergency transportation route network,

3) Upgrading safety of lifeline facilities The policy includes construction of common

utility ducts accommodating water, gas, electricity,

telephone and other pipelines, providing backup

water sources needed for fire fighting and daily

lives, and improving information systems to

collect, analyze, disseminate disaster information.

(3) Technological development

The Public Works Research Institute has been

promoting R&Ds for the urban earthquake

disaster mitigation. Followings are the principal

research themes.

1) Technologies for disaster prevention

- Preparation of seismic hazard map taking effects of inland faults into account

- Evaluation of emergency transportation route seismic performance

Fig.5 Urgent Damage Estimation Technology

(Highway Bridge Damage)

2) Technologies for crisis management

- Urgent damage estimation using ground motion

observation network (Figure 5)

- Urgent damage detection using advanced

information technologies including satellite image

and sensors

33 Flood disasters

(1) Basic concept for disaster prevention

Conventional measures have been promoted in order to prevent flood-induced damage against

anticipated level of rainfalls. Although the frequency of flood disasters has been reduced, the

scale of disasters has tented to increase due to the

urbanization. The basic concept of flood disaster prevention becomes, therefore, an appropriate

control of the relationship between intensive

rainfall and induced damage. Based on this basic

concept, comprehensive flood disaster mitigation measures are needed in order to create watershed

systems where intensive rainfall does not

significantly affect the socio-economic activities.

1) External natural force

Relationship between anticipated intensive

rainfalls and induced damage in the existing

watershed must be carefully studied. Anticipated

form, scale and influence of flood disaster must be

studied considering past disaster experiences,

changes of watershed and safety of related

facilities.

2) Requirement of safety

Requirement of the flood frequency reduction and

flood-induced damage reduction must be studied

considering above described relationships in each

watershed. Combination of disaster mitigation

facilities must be carefully selected considering

the reliability/ deterioration and cost/ effect/

limitation of facilities.

(2) Disaster mitigation measures

423
FoUowings are basic approaches for flood disaster

mitigation implemented by the Ministry of

Construction.

1) Facilities to deal with anticipated floodwater

These include river course improvement, levees,

diversion channels, dams, retarding basins,

drainage pumping stations, high tide levees, etc.

2) Facilities to reduce damage due to extreme

floodwater

These include super levees, forest zones, double

levees, emergency drainage gates, and land side

channel networks.

3) Measures to reduce floodwater

These include the conservation of water retention/

retarding functions by constructing storage/

infiltration facilities in the watershed, and the

control of woody debris and harmful sediment by

constructing sediment dams.

4) Reduction of disaster potential in watershed

This approach includes the management of land

use, and the promoting of flood-proofing city

planning.

5) Improvement of urgent response

This approach includes improvement of urgent

warning, dissemination of hazard map,

improvement of evacuation by education/ drills.

(3) Technological development

Followings are principal research themes

conducted by the Public Works Research Institute

for flood disaster mitigation.

1) Technologies for disaster prevention

- Preparation of flood hazard map for displaying

distribution of risks in each watershed (Figure 6)

2) Technologies for crisis management

- Pre-disaster dissemination of flood information

for the awareness/ education of residents

- Post-disaster dissemination of flood information

for the effective evacuation activity

4. CONCLUDING REMARKS

This report reexamines past earthquake and flood

disasters to study the changes, features and

lessons of urban disasters. This report also

introduced the basic concept, policy measures and

technological development for the urban disaster

mitigation in Japan.

Features of urban disaster can be summarized as

follows;

1) Highly concentrated population and properties

tend to cause an extremely increase of human

casualties and structural damages.

2) Once urban infrastructures lost their functions,

serious repercussions to socio-economic

activities are developed. As the consequence

of highly interdependence of urban functions,

primary damage induces chain-reaction and

Fig. 6 Flood Hazard Map

424
causes the expansion of secondary damage in and around the damaged region.

3) Accompanied with newly appeared urban

facilities/ functions and the chain-reaction

effects, unexpected forms of damage tend to be developed.

Lessons learned from past urban disasters can be

summarized as follows;

1) For securing safety against "low frequency

and high influence" urban disasters, it is

essential to develop appropriate disaster

mitigation plans that include risk management

measures for damage prevention and crisis management measures for minimizing

damage influence. It is also essential to disseminate appropriate disaster information

to public for the better awareness and

education.

2) It is essential to assess the susceptibility of

existing urban regions. Viewpoints include

distribution and attribute of population, newly

appeared facilities/ functions, chain-reactions,

spread of residential districts, advanced use of

urbzm space, and distribution of hazardous

plants and materials.

REFERENCES

1) PWRI: Report on disaster caused by the 1964

Niigata Earthquake, report of PWRI, vol. 125,

1965.6

2) PWRI: Report on disaster caused by the 1978

Miyagi-ken-oki Earthquake, report of PWRI,

vol. 159, 1983.3

3) PWRI: Report on disaster caused by the 1983

Nihonkai-chubu Earthquake, report of PWRI,

vol. 165, 1985.3

4) Earthquake Engineering Division of PWRI:

Report on disaster caused by 1982 Intensive

Rainfall in Nagasaki Pref,, technical

memorandum of PWRI, vol. 2417, 1986.11

5) PWRI: Report on disaster caused by 1995

Hyogo-ken Nanbu Earthquake, report of

PWRI, vol. 196, 1996.3

6) Central Disaster Prevention Council: Basic

Plans for Disaster Prevention, 1997.6

7) Ministry of Construction: Concept of Urban

Planning emphasizing on Seismic Safety,

1995.4

425
Mi
Effects of the August 17, 1999 Kocaeli (Izmit) Earthquake and the

November 12, 1999 Duzce Earthquake, Turkey, on Dams

by

Ellis L. Krinitzsky', Mostafiz R. Chowdhury^ and Ghassan Al-Chaar^

ABSTRACT

Three major dams were situated m areas of

serious ground shaking, >0.2 g, caused by the

1999 Kocaeli and Izmit earthquakes along the

North Anatolian Fault in Turkey. The Kirazdere

Dam was 1 to 2 km from the fault rupture, and the

dam sustained shaking of up to 0.4 g, yet had no

damage. The other dams also were not damaged.

KEYWORDS: Turkey, Kocaeli earthquake, Izmit

earthquake, dams, earthquake damage

1.0 INTRODUCTION

On August 1 7, 1 999, a powerful earthquake struck

Turkey. The shock occurred at 01:39:80

Greenwich Time, 03:01 AM local time. The

magnitude was interpreted at the Kandilli

Observatory and Earthquake Research Institute of. Bogazici University in Istanbul to be Mj:7.8.

Additional values were assigned as follows:

m^ : 6.3 (USGS)

Duration magnitude : 6.7 (Kandilli)

: 7.4 (USGS, Kandilli)

Epicenter : 40.70N, 29.99E (USGS)

Focal Depth : 1 1 km

The epicenter occurred about 11 km SE of

Kocaeli (Izmit) as shown in Figure 1

.

Another strong earthquake occurred at Duzce

(Figure 1) three months later, on November 12. This earthquake occurred with 30 km of surface

rupture on the North Anatolian fault. It was the same fault that ruptured in the Kocaeli earthquake

but its epicenter was about 1 20 km east of Kocaeli

(see Figure 1). Values for the Duzce earthquake

are:

: 7.2 (Kandilli)

M, : 6.3 (USGS)

M, : 7.5 (USGS)

Epicenter : 40.76N, 31.16E

Focal Depth : 10 km

Based on the peak recorded motions, the

distribution of aftershocks, and the extent of

strong damage, isoseismals were interpreted for the Kocaeli earthquake as shown in Figure 1 . The

shaking was experienced as a narrow band

parallel to the North Anatolian fault. Maximum

damage occurred in the central area at Kocaeli

(Izmit) but also in pockets along the north and

south shores of the Gulf of Izmit, particularly at Karamursel and Golcuk. Shoreline property

settled, parts slid into the Gulf, and water

encroached onto the land. To the east, notable

damage from soil liquefaction was experienced at Sakarya (Adapazari). And damage occurred in the west near Yalova.

2.0 BEHAVIOR OF DAMS

Only two major dams were in the zone in which

the isoseismals for the Kocaeli earthquake were

equal to or greater than 0.2 g. The locations of

these dams are shown in Figure 1. They are Kirazdere (Yuvacik) Dam and Gokce Dam.

Kirazdere Dam is within 2 or 3 km of the epicenter and in close proximity to the causative fault. The dam would have experienced peak horizontal

motions of 0.3 to 0.4 g. Gokce Dam is in the area

' Senior Research Scientist, Geotechnical Laboratory, Engineering Research and Development Center, U.S. Army Corps of

.

Engineers, 3909 Halls Ferry Road, Vicksburg, MS 39180-6199, krinite@mail.wes.armv.mil.

^ Research Civil Engineer, Structures Laboratory, Engineering Research and Development Center, U.S. Army Corps of

Engineers, 3909 Halls Ferry Road, Vicksburg, MS 39180-6199, chowdm@ma11. wes. army. mil.

^ Research Civil Engineer, Construction Engineering Research Laboratory, Champaign, IL, g-al-chaar@cecer.armv.mil.

427
of considerable damage and numerous aftershocks

in the area of Yalova. Gokce Dam would have

experienced motions of at least 0.2 g or greater.

This is a reasonable inference though there were

no strong motion records in the Yalova area.

2.1 Gokce Dam

Statistics on Gokce Dam are:

Year completed : 1988

Height (from foundation) : 61.0 m

Height (from river bed) : 50.0 m

Crest length : 747 m

Reservoir area : 1.28 km^

Reservoir volume : 13,250,000 m^

The dam (Figure 2) is composed of a rock-filled

shell having a clay core with upstream and

downstream sand filters. The reservoir level during the earthquake appears to have been at about 70 percent. The dam is founded on

bedrock.

As noted, the dam is interpreted to have

experienced 0.2 g or greater of peak horizontal

acceleration. There were no strong motion

records in the region. The only observed effect ol" the earthquake on the dam was a longitudinal

crack which developed on the crest. The crack is shown in Figure 3 . It was about 7 mm wide at the

middle of the dam and was located along the

upstream side ofthe gravel road which runs along

most of the crest of the dam. The work was

judged by the authors to have been only cosmetic

in its effects. Fine cracks, not affecting the

structure, were observed in a concrete building at the dam.

The downstream face of the dam was observed to be somewhat irregular as seen in Figure 2. The

irregularity is on the order of 1 to 2 m. There was

no evidence that this was the result of slumping or other movement. The exposed upstream face

appears to have a more uniform appearance.

There was also no observed evidence of increased

water seepage downstream from the dam. Except

for the minor crack along the crest, the dam

showed no effects of the earthquake shaking.

2.2 Kirazdere (Yuvacik) Dam

Kirazdere Dam is also known as Yuvacik Dam.

Its statistics are:

Year completed : 1992

Height (from foundation) : 108.5 m

Height (from river bed) : 102.5 m

Crest length : 390 m

Reservoir area ; 1 .74 km^

Reservoir volume : 5,200,000 m"'

Kirazdere Dam (Figure 4) is of particular interest as the epicenter of the Kocaeli earthquake is within a few kilometers of the site. Figure 5 shows two meters of right lateral strike slip fault movement that was observed within one to two

kilometers of Kirazdere Dam. From the

isoseismals in Figure 1, the dam is inferred to have experienced 0.3 to 0.4 g of peak horizontal

acceleration.

The maximum accelerations recorded from the

Izmit station about 10 to 12 km distant were

0.145 g, 0.167 g, and 0.230 in the vertical, northsouth and east-west directions, respectively. To

utilize these data for the Kirazdere Dam, the

acceleration records were scaled to 0.4 g. The

acceleration data points were scaled by factor 1 .74

(0.4 g/0.230 g), where 0.4 g is the desired scale

and 0.23 g is maximum horizontal acceleration

measured. Specfral acceleration curves for the

scaled acceleration records are shown in Figure 6. From this figure, the predominant frequencies are about 0.35, 0.57, and 0.35 Hz in the vertical,

north-south and east-west directions, respectively.

The Kirazdere Dam is earth fill with a clay core, sand and gravel filters, and a rock shell. A road

passes along its crest. The dam manager stated

that a few cracks resulted from the earthquake.

These had the width of about 2 mm and occurred on the gravel road on the crest near the left abutment looking upstream, ^nd were parallel to the axis of the dam. The lengths of these cracks were not observed. The road on the crest is heavily used and the cracks quickly disappeared.

428
A few hairline cracks were seen in a one-story concrete structure at the dam. No cracking was

seen in the concrete of the spillway.

Benches excavated into the left abutment, looking

upstream, showed some falling away of material

from the stepped benches and small talus accumulations of rockfall debris (see Figure 7). The dam manager, explained that some of this material in the talus fell during the earthquake but some of it was there before the earthquake.

The upstream and downstream faces of the dam

were straight and uniform. A tunnel which

carried a water pipeline from under the dam had

experienced a small leak.

The dam appears to have suffered no significant

damage from the earthquake. The cracks, and

talus on the abutment, were exfremely minor.

2.3 Hasanlar Dam

Hasanlar Dam, located about 30 km northeast

of Duzce and 1 5 to 20 km from the fault rupture

in the November 12, 1999 earthquake, is included

here because of similarities to the Kocaeli event.

Statistics for Hasanlar Dam are;

Year completed : 1972

Height (from foundation) : 72.8 m

Height (from river bed) : 70.8 m

Crest length : 67 m

Reservoir area : 2.85 km^

Reservoir volume : 1 ,65 1 ,000 m^

Hasanlar Dam (Figure 8) was, like Gokce and

Kirazdere Dams, built on bedrock with a clay

core, sand and gravel filters, and a rock shell.

The maximum accelerations recorded from the

Duzce station were 0.481 g, 0.312 g, and 0.364 g

in the vertical, north-south and east-west

directions, respectively. To utilize these data on

the Hasanlar Dam', the acceleration records were

scaled to 0.2 g. The acceleration data points were

scaled by factor 0.55 (0.2 g/0.364 g), where 0.2 g

is the desired scale and 0.364 g is maximum

horizontal acceleration measured. Spectral

acceleration curves for the scaled acceleration

records are shown in Figure 9. The predominant

frequencies are about 0.6, 0.35, and 0.65 Hz in the

vertical, north-south and east-west directions,

respectively.

Inspection of the dam showed no effects from

earthquake shaking.

3.0 CONCLUSIONS

Only two dams experienced earthquake shaking

with a PHAiO.2 g. These were Gokce Dam near Yalova, where the PHA was around 0.2 g, and

Kirazdere Dam, southeast of Izmit where the PHA

was as much as 0.4 g. An additional dam,

Hasanlar, experienced about 0.2 g from the Duzce

earthquake. All of the dams are earth

embankments on bedrock. None of the dams had

any significant damage.

The behavior of Kirazdere Dam, which is located

practically at the epicenter of the 1999 Kocaeli

earthquake, is of particular interest. The dam

experienced a world class earthquake of = 7.8. The dam had no damage.

4.0 ACKNOWLEDGMENTS

The authors were greatly helped by Dr. Mustafa

Erdik, chairman of the Department of Earthquake

Engineering and Director of the Kandilli

Observatory and Earthquake Research Institute at Bogazici University, who furnished strong motion

instrumental data and background for

understanding the earthquake processes in the

region. He also provided contacts with Turkish

officials for permissions to visit dams.

The resulting data from this research effort, unless

otherwise noted, were obtained from research

conducted under the Earthquake Engineering

Research Program of the U.S. Army Corps of

Engineers by the U.S. Army Engineer Research

and Development.Center. Permission was granted

by the Chief of Engineers to publish this information. Opinions expressed in this paper

may or may not represent those of the Corps of

Engineers.

429
430
Figure 2. Downstream view of Gokce Dam near Yalova

431
Figure 3. Longitudinal crack near the edge of the crest of Gokce Dam on the upstream side. The location is about at the middle of the dam. The width of the crack was about

7 mm at its greatest

432
433
Right lateral strike-slip fault movement of about two meters which occurred within one to two kilometers from Kirazdere Dam

434
Shock Spectra (5% Damping)

0) >

Vertical (Scaled) N-S (L) (Scaled)

E-W (T) (Scaled)

100

.Frequency (Hz)

Figure 6. Spectral acceleration for the scaled acceleration time history at the Izmit station

developed for Kirazdere Dam

435
Figure 7. Rockfall debris in the left abutment of Kirazdere Dam

436

Shock Specira (5% Damping)

>

Vert - (Scaled)

N-S (L) (Scaled)

E-W(T) (Scaled)

Frequency (Hz)

Figure 9. Spectral acceleration for the scaled acceleration time history at the Duzce Station

developed for Hasanlar Dam

438
DAMAGE OF TRANSPORTATION FACILITY IN THE 1999 KOCAELI AND DUZCE, TURKEY

EARTHQUAKES AND THE 1999 CHI-CHI, TAIWAN EARTHQUAKE

by

Kazuhiko KAWASHIMA'^ and Gaku SHOJI^^

ABSTRACT

This paper describes the damage of bridges in the

1999 Kocaeli and Duzce, Turkey, earthquakes

and the 1999 Chi-Chi, Taiwan, earthquake.

Effect of tectonic ground movement is presented.

Keywords: Seismic Damage, Bridge, Turkey,

Kocaeli Earthquake, Duzce Earthquake, Chi-Chi

Earthquake

1. INTRODUCTION

Extensive structural damage occurred in the

Kocaeli, Turkey, Earthquake (moment

magnitude M^=7.4), August 19, 1999, and the

Chi-Chi, Taiwan Earthquake (M^=7.6),

September 21, 1999. About 17,000 and 2,400

victims were induced in the Kocaeli and Chi-Chi

earthquakes, respectively. In the Turkey, an

earthquake with M^=7.1 (Bogazici

University 1999) occxirred on 12, November,

1999 in Duzce, about 100km east of the Kocaeli

(Duzce earthquake), and about 900 peoples were

kiUed.

In tike three earthquakes, the tectonic ground

movements as well as strong groimd shaking

induced significant structural damage in the

transportation facility. A damage survey was

conducted in Kocaeli earthquake by the first author on September 8 and 9, 1999, and in the

Chi-Chi earthquake by the authors on October 9-

14, 1999. This paper describes the damage of

transportation facility in the Kocaeli and Duzce

eartiiquakes, Turkey, and the Chi-Chi, Taiwan,

earthquakes.

2. KOCAELI, TURKEY EARTHQUAKE

(1) Tectonic Grotmd Movement and Ground

Motion in the Damaged Area

In the Kocaeli earthquake, extensive damage of

transportation facilities occurred in the Kocaeli

and Sakarya region. A part of the right-lateral strike slip Anatolian fault ruptured in east-west

direction for about 100km fi-om Golcuk to Duzce.

As shown in Fig. 1, since the Trans-European

Motorway was almost parallel to the fault, the

fault rupture crossed the Motorway at several

locations. Damage of road facilities was

extensive around the city of Arifiye as shown in Figs. 2 and 3. Arifiye Overpass (No. 3, refer to

Fig. 3) and Sakaiya Center Bridge (No. 5, refer

to Fig. 3) coUapsed. Overpasses near Ariyiye

Overpass (No. 2, 4 and 5, refer to Fig. 3)

suffered minor damage.

Various evidence of right-lateral strike slip

rupture was observed in the area (refer to Fig. 3). For example, a 4.3m offset of a concrete fence of

an automobile factory (Arbas Arac ICadlama

Fabzikasi) was developed at northeast comer

(Point D) as shown in Photo 1. In a poplar grove

at Point G, the largest fault offset measured firom

displaced poplar trees that were originally

planted in Unes was 4.3m, as shown in Photo 2. The feult developed an offset of an about Imdiameter sewer pipe embedded at Point H, as shown in Photo 3. The offset appeared to be on

the order of at least 3m and probably 4m. The

fault rupture continued further west parallel to the Motorway and resulted in a 3.6m ofi&et of a

1.4m-diameter drainage pipe at Point I, as shown

in Photo 4. The Points G, H and I were close to Arifiye Bridge that collapsed in llie eartiiquake.

1) Professor, Graduate School of Science and

Engineering, Tokyo Institute ofTechnology

2) Research Associate, Graduate School of

Science and Engineering, Tokyo Institute of

Technology

439
Damazed Zcie by Kocaeli Earthquake

Under Construction

Bolu Tunnel

Sakarya Viaduct

Ariflye Bridge

q 25km 50km

—I

Fig. 1 Damaged Region in the Kocaeli and Duzce Earthquakes, Turkey

Damage of Rails

650

Fig. 2 Damaged Area near Sapancia Lake

lOkm

I

Poplar Grove No.4

OverBridge

Arbas Fig. 3 Damaged Transportation Facilities and Fault-Induced Ground Movement, Kocaeli Earthquake

440
Photo 1 Offset of a Concrete Fence of 4.3 m at Automobile Factory (Point D, refer to Fig. 3)

Photo 3 Offset of a Sewage Pipe

(Point H, refer to Fig. 3)

Photo 5 Offset of Poplar Trees of 2.5 m near

Tepetarla Station (South side (bottom in Photo) was

about 1 m lower than the north side (top))

Near Tepetarla Station (refer to Fig. 2) where a

large offset of railways occurred, the largest ofEset measured from offset of poplar trees was

2.5m, as shown in Photo 5. As weU as llie right

lateral slip, an offset of ground surface level of

about Im (south side was lower than the north

side) was observed.

Photo 2 Offset of Poplar Grove of 4.3 m

(Point G, refer to Fig. 3)

Photo 6 Collapse of Arifiye Bridge

(Courtesy ofH. Hoashi)

Strong motion accelerations were recorded at various locations by Ministry of Public Works

and Settlement, Bogazici University and Istanbul

Technical University. Fig. 4 shows the ground

accelerations recorded at Sakarya, Adapazeri.

The fault normal acceleration was not recorded

at this site. The peak acceleration of fault parallel

component was 4.07m/s^, and this was the largest

441
s; 0.6^

0.3^ o

2

8

<

^,1 TTfTTT-H--:-

'

EW

Max=-0.42G

^ax=0.26G

10 15

Time (sec)

20 25 2 3

Period (sec)

Fig. 4 Ground Motions Recorded at Sakarya, Adapazeri

^.=70 degree

Fig. 5 Arifiye Bridge

acceleration measured in the Kocaeli earthquake.

It was rather smaller than the amplitude that is generally generated by a M» 7.4 earthquake. The

response acceleration is over Ig at short natural

period, but it decreases sharply as the period

increases.

(2) Collapse ofArifiye Overpass

The Arifiye Overpass was of 4 span simply

supported prestressed concrete bridge (No. 3, refert to Fig. 3). It was supported by two

abutments and 3 wall-type reinforced concrete

piers. It was a skewed bridge with a skew angle

of 65-degree. The substructures and the decks are denoted herein as Al, PI, P2, P3 and A2, and Dl,

D2, D3 and D4, respectively, from north to south

(refer to Fig. 5). As shown in Photo 6, the D2, D3

and D4 dislodged from their supports at south

ends and fell down with the north side still being

442
1.5 H—N*- 9.5m 1.5

\'

( /nur:b j

\

V '/ V '/ V '/

L 3.4 —^Ti— ^

0.1 0.4 <

25.944m

A-A B-B C-C

1| *

|0.05

n

1

r< T-rx

Fig. 6 Cross Section of Arifiye Bridge (Courtesy of F. Inal and H. Hoashi)

1m

*-<t>22

etc 0.25m

8m

1.5m

13.65m

13.65m

Im

\016ctc Im f -^16 etc 0.25m

Fig. 7 Section of Piers, Arifiye Overpass (Courtesy of F. Inal and H. Hoashi)

443
65degree

-0.68m

"S.LG

.\'.LG

-1.88m

0.68m d

LC

A'. 77?

1.88m d.TR

(a) Deformation of Ground (b) Displacement d/^Q

Fig. 8 Effect of Fault Movement

(c) Displacement dj-if

supported by the piers. D 1 rested on ground

almost horizontally.

The decks were of 5 precast concrete U-beams as shown in Fig. 6. They were supported by 5 elastomeric bearings with 300mm x 300mm and

100mm high per substructure. As shown in Fig. 7, the piers were Im thick and 14m wide in longitudinal and transverse directions,

respectively. The design strength of concrete was

30 MPa. This was a standard overpass in the area. The reinforced concrete footing that supported

PI was 5.3m long and 14m wide in longitudinal

and transverse directions, respectively. The

footing of P3 was supported by 8 Im-diameter

cast-in-place reinforced concrete piles. The piles were reinforced by 12 20mm plain bars. The

averaged concrete strength investigated with the

Shumit hammer was 56 MPa and 47 MPa at the

front wall of A2 and the upper surface of PI

footing, respectively.

As shown in Fig. 5, the fault rupture crossed the

bridge between Al and PI with an angle 6 of

about 70-degree. The right-lateral strike-slip fault

forced Al to move in northeast direction with

respect to P1-P3 and A2. If one idealizes a ground rupture as shown in Fig. 8 (a), the ground

displacement at north and south of the fault rupture, di^ and ^5, may be obtained

asdf^ =ds = Do/2, in which is a relative displacement of ground at the fault rupture. Thus,

the groimd movement in transverse and

longitudinal directions of bridge axis may be

written

^N,TR = DQJ2-sm6

dN,w=DQl2'CO&e

at the north of fault, and

^SJ7? = -A)/2-sma

ds,LG ^ -Do / 2- cose

at the south of fault.

Assuming Do "Am and 0 »70-degree, the

ground displacements may be obtained from Eqs.

(1) and (2) as shown in Fig. 8 (b) and (c). The

relative displacement between Al and PI in longitudinal direction, di/jy is 137m, which is much larger than the seat length of about 0.6 m

at Al and 0.45 m at PL It seems that Dl

444
15

10

1 1

Ultimate•

Yield

M =11.7MNm

U

1

1 0.05 0.1 Curvature (1/m)

(a) Momeut-Curvature Relation

0.15

o

1.5

0.5

0

Yield

1 1

Ultimate >

—

=0.035m

=0.533m

u =1.46MN(149tonf)

1

1 0 0.2 0.4 0.6

Lateral Displacement (m)

(b) Lateral Force-Lateral Displacement Relation

Fig. 9 Lateral Force vs. Lateral Displacement of Piers, Arifiye Overpass

dislodged from its supports due to such a large

relative displacement induced by the fault rupture between Al and PI. On the other hand,

tiie foimdations of P2, P3 and A2 were forced to move in south direction. It spears that such a southward movement of foundations resulted in dislodging D2, D3 and D4 from their support at the north ends.

From the reinforcement presented in Fig. 7 and

concrete strength of 30 MPa, the lateral force vs. lateral displacement relation was evaluated based on the standard moment-curvature analysis

(Japan Road Association 1996) as shovm in Fig.

9. The stress-strain relation of concrete developed by Hoshikuma et al (Hoshikuma et al. 1997) was used to represent the conjfinement

effect Assuming that the tributary mass of a pier

is 5800 kN, the response acceleration that resulted in yielding of piers becomes

approximately 0.2 g.

(3) Overpasses around Arifiye Overpass

Near the Arifiye Overpass, there were several overpasses that had the same deck and

substructure type to the Arifiye Overpass. In

general, their seismic performance was good. For

example, the fault crossed a two span simply

supported prestressed concrete overpass (No. 1 Overpass, refer to Fig. 3) at its south abutment,

after crossing the Motorway at about 50 m west

Photo 7 Minor Damage in an Overpass where the Fault Crossed the Overpass at Its Abutment

of the overpass as shown in Photo 7. However,

the damage was minor, i.e., only about 50mmshear deformation of elastomeric bearings

occurred associated with the tectonic groimd

movement.

At a 3 span simply supported overpass (No. 2

Overpass, refer to Fig. 3), located about 400m

cast of No. 1 Overpass, the deck collided with

the north abutment, resulting in crush of a part of

concrete parapet wall. About 20-30mm-shear

deformation of elastomeric bearings occurred

due to the fault movement. A 3 span simply

supported overpass located about 400m west of

the Arifiye Overpass (No. 4 Overpass, refer to

Fig. 3) suffered only minor damage at the parapet

wall of the north abutment, although the fault crossed nearby at the south abutment.

445
Fig. 1 1(a) Groxrnd Accelerations Recorded in the Duzce Earthquake

446
Photo 8 Collapse of Sakaiya Cential Bridge

Fig. 10 Sakarya Central Bridge (Courtesy of

F. Inal. H. Hoashi and K. Kosa)

g

oo<

g

03o

o<

g

TO

(DO

<

Bolu

LMax=0.76G

0.3

-0.3

Bolu

V

Max=0.20G

10 15

Time (s) 20 25 30

2 3

Period (sec)

Fig. 11(a) Ground Accelerations Recorded in the Duzce Earthquake

447
I

0.3 5

©

8 -0.3 o< -0.6

I ^^^^

S 0.6

5^ 0.30

CO8 -0.3 o<

Duzce

T

Max=-0.42G

3

c

o

8

o<

-0.6L

0.6 0.3.

0

-0.3|_'

-0.6

Duzce

V

Max=0.35G

10 15

Time (s) 20 25

2 3

Period (sec)

Fig. 11(b) Ground Accelerations Recorded in the Duzce Earthquake

The short wall-type piers, the short bridge span

and the large constrain of the rigid abutments to the decks contributed to prevent build-up of the

bridge response in the earthquake.

each on both ends tilted on the embankments.

The fault crossed the north abutment.

3. DUZCE, TURKEY EARTHQUAKE

(4) Sakarya Center Bridge

Sakaiya Center Bridge, crossing Sakarya river, on a regional road (No. 5 bridge, refer to Fig. 3)

collapse as shown in Photo 8. The bridge was 8

span simply supported girders with two Gerber

hinges at center. The length was

10m-t^@12m-i-10m=92m. They were supported

by steel piles as shown in Fig. 10. It was built 30-35 years ago. The decks collapsed with a span

Fig. 11 shows ground acceleration recorded at Bolu and Duzce in the Duzce earthquake

(Bogazici University 1999). The peal

acceleration of lateral components was 0.76 g

and 0.82 g at Bolu. Pulses with long durations

are included in the record. Response acceleration

of 0.05 damping ratio was 2.3 g in the Lcomponent. Fig. 12 shows the ground

accelerations recorded at the JMA observatory in the 1995 Hyogo-ken nanbu earthquake.

448
3

co

8

<

3

co

E

o8

o<

3

co

05©

8

o<

0.6

0.3.

0

-0.3.

-0.6

Kobe

NS . Max=-0.83G

Kobe

UD

Max=0.34G

30

2 3

Period (sec)

Fig. 12 Ground Accelerations Recorded at JMA Observatory

in the 1995 Hyogo-ken Nanbu Earthquake

9 Damage of 10-Span Bridge, Duzce

(Courtesy ofH.Hoashi)

Photo 10 Fault that Crossed 10-Span Bridge

(Coiutesy ofH.Hoashi)

449
Photo 1 1 Damage at Expansion Joint and Beanngs

(Courtesy of H. Hoashi)

Acceleration and time are scaled in the same

with Fig. 11. The Kobe observatory record also

had several long-period pulses. The response

acceleration was over 2.5 g at 0.4 second in the

NS (fault normal) component.

A part of the Istanbul-Ankara Motorway that was

under construction sufifered damage (refer to Fig.

1). A 10-span simply supported prestressed concrete bridge with a total deck length of about

2.1 km suffered damage as shown in Photos 9, 10

and 11. The decks were continuous, which

resulted in the continuous behavior of the bridge

in an earthquake. The bridge was nearly

completion when the earthquake occurred. A

fault crossed between two columns resulting in some rotation of the columns. At least two decks were heavily displaced from their original seat.

4. CHI-CHI, TAIWAN EARTHQUAKE

(1) Tectonic Ground Movement and Groimd

Motion

The earthquake was induced by the Che-Long-Pu

Fault. It was a thrust fault, and surface rupture was observed for over 70km between Min-Ju and

Tungshih. Eastem Taiwan is a tongue of the

Philippine Sea tectonic plate that is overthrusting the Eurasia plate.

As shown in Fig. 13, damage occurred in Shihkang region, Tungshih region, Taiping,

Wufeng and Tsaotun region, Nantou region, Chi

Chi region, and Mingchien and Chushan region.

Figs. 14 and 15 show ground accelerations

recorded at Shikhkang (TCU068), Taichimg

(TCU082) and Nantou (TCU076) and their response acceleration of 0.05 damping ratio,

respectively. It is seen that several pulsses with

longer period are included in the ground

accelerations recorded at north. Almost single

pulse with the peak acceleration of 4.9 m/s^ was

included in the EW component at Shikhkang.

Response acceleration of 0.05 damping ratio is about 1.5g at 0.5 second in this record.

Fig. 16 compares the record at Shikhkang

(TCU068) with the ground accelerations

recorded at Bolu (L-component) and Duzce (Lcomponent) in tide Duzce earthquake (refer to Fig.

11), Syhnar parking lot (NS-component) in the

1994 Northridge earthquake and JMA Kobe

observatory in the 1995 Hyogo-ken nanbu

earthquake (refer to Fig. 12). It is noteworthy

that Shikhkang record had larger response

acceleration that the JMA Kobe record at tiie period over 2 second. The record at Bolu has

over 2 g response acceleration at 0.3 second.

In the earthquake, at least eight bridges (BeiFong Bridge, Chang-Geng Great Bridge, ShcWci Bridge, Yi-Jian Bridge, Wu-Shi Bridge, PinLing Bridge, Mou-Loh-Shi Bridge, Min-Ju Great

Bridge, and Tong-Tou Bridge) collapsed, and a number of bridges suffered extensive damage.

The tectonic movement along the feult developed

most extensive damage. Two bridges

significantly damaged by the fault movement and

a modem cable-staywi bridge damaged by

ground motion effect are described below.

450
Root 1

Roots

Fengyuani

' Bei-Fong Br. /

^Chang-Geng Great Br. 'shc-WciBr.

Don-Fong Br. iTungshih

Taichung

Chan^ua

Tali

(ufeng

Yi-Jian Br.

Che-Long-Pu Fault

Yen-Fong Br.

Tsaotun

Nantou

Mingchien

Min-Ju Great Br.

Wu-Shi Br.

Mou-1

Pin-Ling Br. 'Nan-Gang Br. Rui-Mei Br. J ^hungliao

1 Chichi

igBr.

Mocm Lake

fen-Pin Br.

.Shuili

Chi-DaBr.

'Chi-ChiBr. Ling-Mei Great Br. Ling-Mei Br.

JTong-Tou Br. 10km 20km

Fig. 13 Damage Bridges, Chi-Chi Eardiquake

451
3

co

8

o<

co

to©

8

o<

c

o

CDw

(D

0.6

p 0.3 - 0

-0.3 - -0.6 - 0.6

0.3 - 0 - -0.3 . -0.6 - 0.6

0.30

-0.3 -0.6

0.3 r-

-0.3 0.30

-0.3 0.30

-0.3

Shikhkang (TCU068)

NS

Max=-0.37G

Shikhkang (TCU068)

EW

Max=0.5lG

Shikhkang (TCU068)

Max=0.53G

J I 1 I I I I L. 10 20 30

Time (s) (a) Shikhkang (TCU068)

40 50

Taichung (TCU082)

NS

Max=0.19G

Taichung (TCU082)

EW

Max=-0.23G

Taichung (TCU082)

UD

Max=-0.13G

10 20 30

Time (s) (b) Taichung (TCU082)

40 50

Nantou (TCU076)

NS

Max=0.43G

Nantou (TCU076)

EW

Max=-0.35G

30

Time (s) (c) Nantou (TCU076)

Fig. 14 Ground Accelerations, Chi-Chi Earthquake

452
coS

o

8

<

e. -T 1 1 mo

1.5 -'EW _ UD

1

« •

• « ^ =0.05

0.5 J V _ "•r - - » _

-"•---».,

0 1 i 1 0 1 2 3 4 5

Period (sec)

(a) Shikhkang (TCU068)

2 3

Period (sec)

(b) Taichung (TCU082)

NS

EW

UD

I =0.05

''^"Vr

2 3 4 5

Period (sec)

(c) Nantou (TCU076)

Fig. 15 Response Accelerations of 0.05 Damping

Ratio, Chi-Chi Earthquake

(2) Coll^se ofBei-Fong Bridge

Bei-Fong Bridge was located downstream of the

Shikhkang Dam, which suffered extensive

damage by the feult movement. Due to the fault movement, a waterfell was created about 100m

upstream of the bridge. It was of 13-span simply

supported I-beam girder, constructed in 1991.

The south most three spans collapsed as shown

in Fig. 17 with other spans being free from

damage. The abutments, piers and decks are numbered herein from north to south. A2, P12

and PI 1 were upheaved about 3-4m, and P12 and

A2 were laterally displaced about 3.5m and 4m,

respectively, in downstream direction (west) as shown in Photo 12. It is known that the reversed

fault accompanied lateral displacement at many

locations. D13 was resting on the slope of river

embankment with the north end being dislodged

6.6m south from the original position.

453
0.9^

0 10 20 30 40 50

Time (sec)

Ol I I I I ' ^ * —

0 1 2 3 4 5

Period (sec)

Fig. 16 Comparison of Strong Ground Accelerations in the Duzce, Chi-Chi,

Hyogo-ken Nanbu and Northridge Earthquakes

4

454
Photo 12 Lateral Movement ofA2 and P12,

Bei-Fqng Bridge

c ^ w VT About 25m

D13 D12 Dll DIP | < >| Dl

P12Tilted PllOvertumed About 10m Al

3~4mUpheavai

D13 /£!? Dll DIO

A2 P12 ^dtTT*

About 10m

(a) Side View from East

S-< •N jlOm Dl

10

Fault N20E~N40E

PI Al

(b) Plane

Fig. 17 Damage ofBei-Fong Bridge

Bei-Fong Bridge

(j lyOm

Taiwan Geological Sxirvay

K.Otsuki (Tohoku University)

Fig. 18 Fault aroimd Bei-Fong Bridge

Lateral Movement Upheaval

Fig. 19 Fault Movement and Ground Movement,

Bei-Fong Bridge

455
Fig.20 Shortening of Span Length due to Upheaval

North

Fig. 21 Collapse Mechanism, Bei-Fong Bridge

Fig. 18 shows the location of fault investigated

by the Taiwan Geological Survey and Otsuki, K.,

Tohoku University. Probably the fault crossed

the bridge between A2 and P12 with an angle of

about 42-degree. From the fact that PI 2 was

laterally displaced in west, and that Dl 1 and D12

were clockwise rotated (20-40-degree), it appears

that tiie fault movement crossed the bridge in

wider zone; perhaps between A2 and Pll as shown in Fig. 19.

If one idealizes tiie fault movement at south of

the faxilt relative to the north, the transverse and

longitudinal displacements of ground may be

represented as (refer to Fig. 19)

456
dxR = -A)Sin0

(3)

duG —DQOOse

in which Db is a fauh displacement

Substituting ^77? »-4m and 6 «»42-degree mto

Eq. (3), one obtains Dq —6m and du^ - -4.5m.

The longitudinal displacement dix} •—4.5m is shorter than the amount of above-mentioned

southward dislodgment of D13 at the north end

(6.6m), however they are not too much different

It is interesting to note that the vertical upheaval

might not be a serious cause of the dislodgment

of decks from their supports. If one idealizes the

upheaval at right substructure relative to the left as shown in Fig. 20, tiie shortening of span

lengtii due to the upheaval. A/, may be written

as A/ L-(i_cos0v) (4)

COS0y

in which 0 v is angle oftilting, and is given as 0v = tan-^(y-) (5)

in which / is span length and dy is amplitude

of relative upheaval between two substructures. Assinning A/ = 25m and dy •=>3m, one obtains

0y=6.% degree and A/ =0.1 8m. Consequently,

shortening of the span length due to upheaval is shorter than the seat length.

Based on the above facts, it spears that the

failure of the bridge was triggered by a large

tilting of Pll in north direction, caused by fault movement as shown in Fig. 21. D12 dislodged

first from Pll at its north end, followed by the

dislodgment at south end. This resulted in the

complete collapse of D12. D12 finally feh down

on the sidewall of Pll at north end with the

clockwise rotation associated with right-lateral fault movement The collapse of Pll resulted in the dislodgment of Dll at its south end. The

south end of Dll thus finally felt down on the

north end of D12. Increased distance between

PI 2 and A2 resulted in the collapse ofD13.

(3) Collapse ofWu-Shi Bridge

Wu-Shi Bridge was of a separated northboxmd

bridge (upstream, east) and a southboimd bridge

(downstream, west) as shown in Fig. 22. They

were of 17-span simply supported PC beam

girders. Bridge axis was N20E. The abutments,

piers and decks are numbered herem from north

to south, and the northbound and southbound

bridges are designated by putting E and W,

respectively.

The northbound bridge was constructed in 1983,

while the southbound bridge was newer than the

northbound bridge. As shown in Photo 13, the

northbound bridge was supported by wall-type

piers with a section of 8.5m x 3m, while tiie southbound bridge was supported by reinforced concrete piers with a rectangular section of 5m x

2m. For example, in PI W (PI at the southbound

bridge), 32 deformed bars with a diameter of

22mm were placed around the pier, and

deformed tie bars with a diameter of 12mm were

placed 200mm interval. They were supported by

caisson foimdations with a diameter of 6m.

Deformed 22mm bars from the caissons were

spliced with the bars in the piers within Im from

the bottom of piers. Concrete strength measured

by the Shumit hammer was about 24MPa.

DIE and D2E collapsed (refer to Fig. 22), and

DIW and D2W tilted and settled extensively. Fig.

23 shows damage of piers in the southbound

bridge. PIW and P2W failed in shear, as a result of a shear force from east to west The caisson

foxmdation at P3W suffered shear failure due to a

shear from east to west as shown in Photo 14. Damage was significant at PIW and P2W,

however other piers also (P3W-P8W, P13W)

suffered damage.

On the other hand, damage of P2E and P3E was

limited; P2E suffered several shear cracks at midheight as a result of lateral force from north

to south. P3E suffered a lOO-lSOmm opening of

concrete at 1.5m high from the foimdation. P2E

and P3E looked as if tiiey settled about 2m

relative to tiie piers in the south beyond P4E. It seems that this was induced by the upheaval in the south ofthe fault. A 2m high scarp was observed at about lOOm

east (upstream) of the bridge. The fault crossed

the bridge between P2 (P2E & P2W) and P3

(P3E & P3W) with an angle of about 40-degree

457
DIE, D2E

Collapse

N20E

About 30m

Northbound Bridge

l.SmUpheaval

AlE PIE P2E P3E —or

P4E

tr

P5E

D17E

About 10m P16E

A2E

DIW D2W D3W

Decks Tilted and Settled About 30m

Southbound Bridge

-fn—a—Q—a—i^ia—ci

n—cr PIW P2W P3W P5W ' P8W Abou\lOm

Shear Failure Shear Cracks P6W, P7W Shear Cracks tjhw

Flexural Cracks Liquefaction

(a) Side View from West

N20E Northbound Bridge

* t » XZZEZZEZZr XI

DIE, D2E

Collapse D4W, D5W Southbound Bridge

I II \ I \ » * i -r X XI

DIW, D2W

Decks TiltedFault N60E (b) Plane

Fig.22 Damage of Wu-Shi Bridge

Photo 13 Collapse ofWu-Shi Bridge

(N60E) as shown in Fig. 24. Between P2 and P3, a vertical rupture of about Im high (piers in south beyond P3 were upheaved) was observed.

Photo 15 shows a displacement of PIW and P2W

in east direction (transverse direction), relative to other piers in south beyond P3. The displacement was as large as 2rn. However it was the ground at south of the fault (between P2 and P3) that moved in the earthquake. Photo 15 should be

consequently interpreted that piers at south

beyond P3W were forced to be upheaved

resulting in the lateral displacement in west

direction. It appears that this attributed to the fact that not only PI and P2, but also most piers from

P3 to P16 suffered damage.

Substituting the relative displacement in transverse direction of djj^ » 2m and 6 -•40- degree into Eq. (3), the fault dislocation may be

DQ'=>3.lm, and relative displacement in longitudinal direction djjc is about 2.4m. It appears that the westward fault movement at south beyond P3W resulted in a lateral force to occur in DIW and D2W in west direction, which,

in turn, caused the shear failure in PI W and P2W.

It is interesting to note that since the failure of

PIW and P2W contributed to absorb the large

fault displacement, the decks (DIW and D2W)

did not collapse.

In a similar way, a large lateral force appUed to DIE and D2E in the northbound bridge, which

resulted in some damage at PIE and P2E. Since

strength of wall-type piers (PIE and P2E) was

large enough, they did not collapse. However,

458
North Side South Side L ^ L I

P8W / 1

P7W

1 P6W [

/ I P5W [

I 1 P4W [

i 1 P3W I .^^^i

-A—A A Taiwan Geological Survay Not Confirmed

> K. Otsuki (TcAoku University) ^ Kawashima, Shoji and Iwata

Fig.24 Fatilt around Wu-Shi Bridge

Photo 14 Failure of Caisson Foundation as a

Direct Fault Movement, Wu-Shi Bridge

(Courtesy of S. Iwata)

Fig. 23 Failure ofPiers, Wu-Shi Bridge

459
Photo 15 Lateral Movement ofPI and P2,

Wu-Slii Bridge

this developed large relative displacements

between the piers and the decks, and this resulted

in tiie collapse of DIE and D2E.

It appears that the failure of the caisson

foundation at P3W (refer to Photo 14) was

developed by a direct fauh movement.

(4) Damage ofChi-Da Bridge

A cable-stayed bridge as shown in Fig. 25

suffered damage at the tower, deck and cables. It was a single towered two-span continuous

prestressed concrete cable-stayed bridge, and it was under construction. The tower and the decks

460
were rigidly connected. Multi-cable (17 cables)

in double planes was used.

The tower suffered spalling ofif of covering

concrete as shown in Photo 16. Since the damage

was more extensive on the surface in transverse

direction, it speared that the damage was

developed associated with the tower response in transverse direction. A 100-300mm wide flexural cracks penetrated the upper and lower slabs of

the deck at close to the tower as shown in Photo

1 7. It seemed that the flexural crack occurred due

to longitudinal response of the bridge. The 7th

cable from bottom suffered damage at its cast- iron coupler, and was pulled out It appeared that several other cables suffered damage to some

extent at the connections. No evidence of a fault movement was observed around the bridge.

5. CONCLUSIONS

Seismic damage of bridges in the 1999 Kocaeli

and Duzce, Turicey, earthquakes, and the 1999

Chi-Chi, Taiwan, earthquake was presented. Features of the damage may be summarized as follows:

(1) TTie tectonic ground movement that crossed

bridges resulted in the extensive damage in Kocaeli, Duzce and Chi-Chi earthquakes.

(2) In Kocaeli Earthquake, Arifiye Overpass

collapsed due to a fault movement that crossed

the bridge with an angle of 65-degree. An

increased distance between Al and PI resulted in

tiie collie of Dl, and the lateral displacement

of foundations forced to move by the fault contributed to the damage of D2, D3 and D4.

Because other overpasses on the Trans-Europe

Motorway were rested on short and wall-type

piers and abutments with large cross sections,

damage due to the ground motion was less significant

(3) In Duzce earthquake, a new and long

prestressed concrete bridge supported by high

piers suffered damage at their supports. A fault movement that crossed the bridge was observed,

and this may be the main reason for damage.

(4) In Chi-Chi Earthquake, at least eight bridges

collapsed. Bei-Fong Bridge and Wu-Shi Bridge

suffered damage associated with the fault

movement The fault crossed the bridges that resulted in a large lateral force in the piers.

Overturning and failure of the piers triggerai the

collapse ofthe bridges.

(5). It is interesting in Wu-Shi Bridge to note that the decks supported by stiff wall-type reinforced concrete piers collapsed in the northbound bridge

while the decks supported by reinforced concrete

piers with smaller rectangular sections did not

fall down (although they suffered major damage)

in the southbound bridge. Failure of the piers in the southbound bridge contributed to absorb a

large relative displacement induced by the fault movement and as a result, it prevented to have a

critical failure. It shows that a certain type of

devices or mechanisms that absorb the fault displacement are effective to prevent failure in a

bridge subjected to a large fault movement.

(6) A modem cable-stayed bridge suffered

extensive damage at the tower and the deck.

Since the tower and the decks were simply fixed,

this may be a major contribution to the damage

ACKNOWLEDGEMENT

The authors express their sincere appreciation to Fulya Inal and Hiroaki Hoashi, Division of

Motorway Bridges, General Directorate of

Highways, Ministry of Public Works and

Settlements, Turkey for their kind cooperation

for providing valuable information. The authors

also sincerely appreciate Professor C. H. Loh and

K. C. Chang, Center for Earthquake Engineering

Research, National Taiwan University, Taiwan

for their kind support for the site damage

investigation. Kind cooperation of Takao

Hashimoto, Chiyoda Consultants, and Takeo

Suzuki, Okumura Construction, for the site investigation in Kocaeli Earthquake, and

Professor Kazuhiro lemura, Kyoto University

and Shuji Iwata, JR Tokai, for the site investigation in Chi-Chi Earthquake are greatly

appreciated. Valuable information for Kocaeli

and Chi-Chi earthquakes was provided by

Professor Kenji Kosa, Kyushu Technical

University.

461
REFERENCES

Bogazici University Kandilli Observatory and

Earthquake Engineering, Home Page, 1999

Hoshikuma, J., Kawashima, K., Nagaya, K. and

Taylor, A., Stress-Strain Model for Confined

Reinforced Concrete in Bridge Piers. Journal

of Structural Engineering, 123(5), 624-

633.New York: ASCE, 1994.

Japan Road Association, Part V Seismic Design,

Design Specifications of Highway Bridges.

Tokyo:Maruzen, 1996.

Kawashima, K., Hashimoto, T. and Suzuki, T., Damage of Transportation Facilities in the

Kocaeli, Turkey, Earthquake. Report No.

TIT/EERG 99-7, Tokyo Institute of

Technology, Tokyo, 1999.

Kawashima, K., lemura, H., Shoji, G. and Iwata,

S., Damage of Transportation Facilities in the Chi-Chi, Taiwan Earthquake. Report No.

TTT/EERG 99-8, Tokyo Institute of

Technology, Tokyo, 1999.

Lee, W. et al, CWB Free-field Strong Motion

Data from the 921 Chi-Chi Earthquake, Vol.

1, Digital Acceleration Files on CD-ROM,

1999

USGS, Implications for Earthquake Risk

Reduction in the United States from the

Kocaeli, Turkey, Earthquake of August 17,

1999, Geological Survey Circular, 1999

462
The Performance of theTrans European Motorway Structures

During the Nov. 12, 1999 Diizce Earthquake

By

Hamid Ghasemi', James D. Cooper^ Roy Imbsen^

ABSTRACT

Almost 3 months after the devastating August

17, 1999 Kocaeli earthquake, another earthquake

with a moment magnitude of 7.2 hit Turkey on November 12, 1999. Called the Duzce

earthquake, it resulted from a right lateral strikeslip rupture along the secondary Diizce fault. Its epicenter was centered near the town of Diizce,

in the province of Bolu. This earthquake caused

close to 1000 fatalities and 5000 injuries. The

damage to buildings was similar to that sustained during the Kocaeli earthquake. Two

viaducts and one tunnel under construction

exhibited extensive damage.

The content of this paper is based on a reconnaissance survey, of bridge and tunnel sites along the Trans European Motorway (TEM)

segment under construction near Bolu,

conducted by the FHWA team dispatched to Turkey on November 28, 1999. During this survey, the roadway alignment and the physical

condition of the Bolu Viaducts #1 and #2, and

Bolu Bridge #2 were observed; superstructure

damage to Viaduct # 1 was viewed from the east and west abutments and from atop Piers #4, #5,

and #10 (the first expansion joint from the

western abutment of the westbound span); Bolu

Bridge #2 damage was viewed at the west

abutment seat and from atop the first pier east of

the western abutment of the eastbound span. Bolu Tunnel damage was viewed from the

Elmalik entrance westward to the collapse in both tunnels.

KEYWORDS: Duzce EQ, Kocaeli EQ, fault rupture, near fault effect, energy dissipating

units, bearing failure, tunnel collapse

1.0 INTRODUCTION

The November 12, 1999 Duzce earthquake with

a moment magnitude of 7.2 was caused by a

right lateral strike-slip rupture along the

secondary Diizce fault near the town of Diizce

(figure I). This fault is connected to the main

segment of the North Anatolian Fault (NAF) by

the Elmalik and Asagi Bakacak faults. The

length of the surface fault rupture is estimated to be 40 km with an average lateral offset of 4 m

along most of its length.

The NAF has been studied as a collaborative

effort by both Turkish and U.S. scientists for many years. In particular, U.S scientists have

been interested in the NAF because there is a

strong similarity between the creep rate and

energy release of both the North Anatolian and

the San Andreas fault in California (figure 2). Both faults have generated large magnitude

earthquakes in the last 100 years.

Figure 3 shows the close proximity of the TEM

to the NAFZ which is of outmost importance

and depicts the epicenter of the August 17, 1999

Kocaeli Earthquake, located in the town of

Golcuk, about 140 km east of Diizce.

'Ph.D., Research Structural Engineer, FHWA, TumerFairbank Highway Research Center, McLean, VA

^P.E., Technical Director, Bridge R&D, FHWA,

Tumer-Fairbank Highway Research Center, McLean, VA

^Ph.D., P.E., D. Engr., President, IMBSEN &

ASSOCIATES, INC., Sacramento, CA

463
0 Nov. 12, 1999 Duzce EQ (M^ = 7.2) Recent Fault Rupture

-100 0 100 200

Distance east of 35'£ longitude (km)

Figure 1: The North Anatolian Fault Zone

(;ailh4uaWL0 t>e<tuor>cn -;

F.-mU Slip Ftatc SAt4 mmJyt

Ankmn ^ 'SJ!* 1%I4 /Voflwfsl. T U n K K Y

500 ki«

fnuit Sttp flow CA LIFORNIA

Similar slip rale. age. length and stmightness

Ros> S. Stcia

Figure 2: The Comparison of the North

At the Diizce station near the epicenter of the

Nov. 12 earthquake, Peak Ground Acceleration

(PGA) of 1 -Og was recorded before the ground

motion instrumentation was clipped due to its limitation on recording acceleration above 1 .Og.

Between these two stations, in the towns of

Kaynasli and Bolu, there were two viaducts and one tunnel under construction. It is therefore

reasonable to assume that these structures, that were not instrumented, experienced a PGA of at

least 0.8g. These structures are part ofthe last 24 km segment of the Trans European

Motorway (TEM) that is scheduled to be

completed.

The Diizce earthquake caused considerable

damage to the superstructure of Viaduct #1 and

to the Bolu tunnel due to the close proximity of

tolian and San Andreas Faults

fault rupture (near fault effect). A surface fault trace was visible between a segment of

Viaduct #1 piers and evidence of high velocity

impulses was observed from the earthquake

records and at the sites.

There are indications that high velocity impulses

or "Fling" effects normal to the direction of

rupture of strike-slip faults are produced by near

field effects of earthquakes which present

special problems for highway structures and tall buildings. These problems are manifested by

very large displacements, overturning moments,

the occurrence of liquefaction, and other energy

sensitive structural responses.

464
Figure 3: Map showing approximate proximity of the NAFZ to the TEM

2.0 SITES VISITED

2.1 Bolu Viaduct #1

Viaduct #1 with its 59 dual spans and 2.3 km

structure length was approximately 95%

complete and awaiting installation of expansion

joints for completion of the project at the time of

the Diizce earthquake (figure 4). Its span are 40 m long and consist of 7 lines of simplysupported, prestressed concrete box girders

seated on pot bearings with a stainless steel PTFE-slider interface. The deck slab is continuous over 10 spans.

The viaduct's piers are single, octagonal cast-inplace, hollow-core reinforced concrete columns,

4.5x8.0 meters in plan dimension with heights

varying from 10 m to about 49 m. They were

designed and detailed to provide ductile

behavior during earthquakes.

The viaduct had also incorporated an Energy

Dissipation Unit (EDU) system, which is installed on each pier cap to accommodate

longitudinal thermal movements and to reduce

any seismic forces through energy dissipation

during a major event.

Figure 4: General View of Bolu Viaduct #1

2.1.1 Design Strategy

The design concept for Viaduct #1 involved the

application of mixed criteria, using the 1990

AASHTO seismic design guidelines coupled

with the design of a hybrid isolation system

based on Italian guidelines. The system is comprised of pot bearings with a Teflonstainless steel sliding interface under each boxgirder and an EDU located at the center of each

pier-cap which was bolted to a high reinforced concrete pedestal on top of the pier cap and to

465
the underside of the deck (figure 5). The

arrangement of EDU's for a 10-span simply

supported segment of the viaduct is shown in the

Figure 6. Also as shown in Figure 6, the deck

slab is continuous over the intermediate

supports.

The EDU's consist of C-shaped energy

dissipating steel elements which are referred to as crescent moons (figure 5). These elements

provide hysteretic behavior through yielding of

the steel elements. In addition to the crescent moons, a piston and a sliding unit are incorporated into the EDU's at the expansion

joints and at the center pier of each 10-span

continuous segment.

Cable restrainers were utilized at the expansion

joints, (figure 7), as a secondary line of defense

to prevent the end girders from falling off their

supports due to displacements beyond those

assumed.

In summary, during normal environmental

conditions (i.e., thermal expansion) the viaduct

movement is provided for by the PTFE-slider

interface of the bearing in conjunction with the

sliding unit of the EDU's which is controlled by

the piston and is incorporated into the units. However, during the design level earthquake

(acceleration of 0.4g), the piston would lock up

similarly to a shock absorber and engage all of

the EDU's on each 1 0-span continuous deck

segment. This would dissipate the energy

induced by the ground motion and also reduce

the displacement response and the total force

exerted on each pier.

Figure 5: A Pot bearing and an EDU unit

466
E = EDU P = Piston R = Restrainers

Figure 6: The arrangement of EDU's with and without pistons and cable restrainers

300000 -r

figure 7: Cable restrainers

at the expansion joint

Figure 8 shows a typical force-displacement

characteristic of an EDU at the ultimate

displacement (±480 mm). The enclosed area is the energy dissipation capability of the unit

which was designed for 40% of critical damping.

The team believed that a good seismic design

philosophy had been selected with an

apparent ductile reserve capacity of the concrete

piers for resisting earthquakes beyond the design

level earthquake.

Figure 8- Actual testing result of an EDU at ultimate displacement (±480 mm)

2.1.2 Field Observations

1 ) The surface fault trace was visible

sporadically in the valley town of Kaynasli,

where Viaduct #1 is located. Faulting caused

major destruction to buildings, residential

housing and other facilities. The fault trace was

also observed at Viaduct #1, where it intersected

Pier #45(eastbound), then continued between the

two Piers #46 before intersecting Pier #47

(westbound) as shown in Figure 9. Fault rupture

movement, estimated to be about 2.5 to 3 m at

this location resulted in a rotation of nearly

467
12 degrees at the pile caps of piers #45

(eastbound) and #47 (westbound) about their

vertical axes (figures 9 and 10).

2) Significant damage did occur to the EDU's

and bearings from surface rupture displacements

and ground shaking which caused the

superstructure girders to translate on their pier

tables, narrowly avoiding total collapse, (figures

11 through 13).

Figure 1 1 : Dislodged EDU, Pier .5

''i'M.'

Figure 12: Typical dislodgment of

bearing systems

Figure 13: End of facia girder extending

beyond pier table

468
3) Shear keys or blocks used to control transverse girder movement, although severely

damaged, appear to have functioned as intended

(figures 14 and 15).

2.1.3 Damage Scenario

The following outline presents a damage

scenario for Viaduct #1 . It is based on the field observations made on November 30, 1999 and

December 2, 1999.

1 ) The viaduct was first subjected to a low

frequency vibratory motion, which was then

followed by a high velocity impulse or "fling"

resulting from the near field effect. The surface

fault trace was visible at Piers 45, 46 and 47

(figure 10).

2) The fling caused the bearings to be ejected

from underneath the box girders after the

bearing sliding plate interface area was exceeded

(figure 12).

3) On some piers, especially those closer to the

fault trace, the fling broke the epoxy bond of the

bearing plates to both pier caps and the bottom

flange of the box girders. Many of the bearing

plates were found on the ground below.

4) On other piers, the stainless steel interface

plates, which were epoxied to the bottom

bearing plates, were protruding from the edge of

the pier cap or were found on the ground below.

5) Almost all of the stainless steel interface

plates had a distinct scoring which resembled the

number "9". The "9" trace may validate a

scenario of a large cyclic motion resulting from

the fling effect (figure 16). Also included in Figure 16 are the authors suggestions for

469
remediation which was made at the time of the

site investigation.

6) Upon the loss of the bearings from

underneath the girders (and in some instances

the steel bearing plates too), the superstructure

then dropped approximately 5 cm to rest on the

pier caps. That caused an excessive bending

moment at the EDU connection to the deck,

which resulted in the partial to total binding of

the C-elements ring plates {figures 17 through

19). The binding action prevented the EDU's

from functioning properly.

7) The fling effect possibly sheared off the

connection between the EDU and its pier cap as well as the connection to the deck. It is believed

that the design of the EDU connections to the

Super- and sub-structure was based on the

elastic limits of the C-element. The connections

for the bearings plates were designed for a force

level equal to the dynamic coefficient of friction

of the PTFE-sliding bearing (4%).

8) It was also noted that the displacement

capacity of the bearings (+2 1 0 mm) was less than those of the EDU units (+320 mm).

Therefore, the EDU's would have never reached

their optimum performance.

9) Most of the 1 0-span superstructure

segments moved toward the west abutment. At

the 1 0-span unit which included piers 45, 46,

and 47 where the fault trace was visible, three of

the girders slid off the cap of Pier # 46 (figure

13). This was possibly the result of rigid-body

substructure motion to the east. The

superstructure rested on PTFE-sliders, had a

greater inertia! mass relative to the substructure,

and therefore was unable to follow the pier's

movement. At the same time, the EDU's were

probably not functional after the initial velocity

impulse, and were unable to control the excess

superstructure displacement. Whether the piers,

the deck, or both moved can be better

determined after a final and precise survey ofthe

bridge is made.

1 0) It appeared that the piers were not subjected

to many flexural cycles of motion as indicated

by the lack of flexural cracking in the pier. The

only hairlines flexural cracking noted were at Piers #35 and #36. The sudden failure of the

bearings and EDU's apparently uncoupled the

substructure from the deck's inertial force. This

fuse type behavior points out the effectiveness of

an isolation approach in protecting substructure

elements.

470
750 mm

Suggested replacement of bearing and slider plates to match expected design

displacement of EDU units

Typical "9" trace on

jdisplaced/ejected bearing plates after earthquake

Viaduct #1 Damage Scenario

from Diizce Earthquake

• Vibratory induced superstructure motions during primary, longer period

motion of earthquake followed by

large offsets from a fling type, high energy input. High frequency spike of

acceleration broke the bond between

the bearing plates. A combination of

pot bearing movement on the plate and dragging of plate propelled both

bearing plates and pot bearings from

underneath the box girder.

• EDU's probably did not perform

as anticipated because offsets from

high energy, fling type motion broke EDU connection to superstructure and

pier table.

• The bearings did not perform as designed because of limited stainlesssteel plate dimensions for the interface

surface.

PTFE Performance May

Not Have Been

Fully Functional For Sliding

Stainless Steel Slider Plate

Bearing Plate

Expand to 1000 mm in redesign, but check pot bearing / EDU

system performance for maximum credible earthquake

Figure 16: Viaduct #1 Bearing Damage Scenario (based on the site review on 1 1/30/99)

471
Figure 17: Bearing failure Figure 18: EDU failure

2.2 BOLU BRIDGE #2

Bridge #2 is an in-service, dual 5-span simplysupported prestressed concrete box girder

structure having a circular pier column and a very large horizontal radius of curvature (figure

20). The spans sit on rectangular reinforced

elastomeric bearing pads and EDU's similar to those used in Viaduct #1 to limit seismic

response. The deck slab is continuous over 5 spans.

2.2.1 Field Observations

1 ) Moderate structural damage occurred to the

concrete box girders of Bridge 2. The west end

of the eastbound span translated approximately

3-1/2 inches to the north relative to the west

abutment, as evidenced by the misalignment of

lane striping at the west abutment, (figure 21).

2) Examination at the abutment seat revealed a

set of well proportioned steel reinforced

rectangular elastomeric bearing pads, (figure

22), and an EDU device that had "worked"

during the Diizce Earthquake,(yzgwe 23). One

"C" element had compressed to a 23-1/2 inch

opening from center of pin to center of pin while

the other "C" element extended to an opening of

3 1 inches. Markings on a bearing plate

472
Figure 21 : Translation of the eastbound span

at the west abutment

indicated that the 12 inch by 27-1/2 inch by 4-

1/2 inch thick bearing pad had translated about

6-1/2 inches to the north and about 3-3/4 inches

to the west. Movements of the elastomeric

bearings allowed the EDU to function and

dissipate energy as conceptually designed.

3) Inspection of the eastbound span at the first pier table to the east of the western abutment

indicated that the bearings allowed the EDU to "work". It was noted that one of the anchorages

of the EDU brackets that was attached to the

superstructure appeared to be at incipient failure, that is cracking of the concrete at the EDU

anchorage point was noted, (figure 24).

473
Additionally, the grade of the superstructure

resulted in a natural 2 inch differential offset between the bracket attachment between spans. This differential distance provides a naturally

occurring, biased moment or twisting to occur

across the EDU which could bind the top and

bottom rings, thus locking the "C" elements and

rendering the EDU inoperative.

Figure 22: Well proportioned steel-reinforced

elastomeric bearing pad showing

evidence to translation

Figure 23: Flaked-off paint of an EDU "C"

element that had "worked"

Figure 24: Incipient failure of EDU anchorage point

474
4) The stable performance of the well

proportioned steel reinforced elastomeric bearing

pads and bearing plates having sufficient sliding

capacity in, all likelihood, allowed the EDU to function, although any additional shaking could

have failed the positive attachment of the EDU

top brackets to the superstructure. When using

isolation/EDU systems, it is imperative to design

the independent units (bearings, plates and EDU)

to act as a SYSTEM.

2.3 BOLU TUNNELS

At the time of the Duzce earthquake, excavation

for 3 .3 km twin Bolu Tunnels was about 70%

complete with about 30% of the final liner system in place. The Tunnels are located about

35 km east of the epicenter. They both have a width of 12 m, outer and inner heights of 15.3 m

and 8.6 m respectively, and outside diameter of

17 m. The bore hole information, field observations and the study of regional geology,

reveals that for approximately the first 1 .5 km

from the Elmalik portal a very soft compressible,

clayey type of material exists. The remaining

1 .7 km is located in weak to moderately

competent rocks. It is also noted that the rock

portion of the tunnels is heavily deformed by

vertical faults suggesting an existence of a wide

brittle shear zone. Figures 25 and 26 show the

tunnel portals at Elmalik (east-bound) and at Asarsuyu (west-bound).

2.3.1 Design Approach

The Bolu Tunnels design methodology is based on the New Austrian Tunneling Method

(NATM). Each tunnel consists of three layers

(from outside inwards) of shotcrete, reinforced concrete and finally an unreinforced concrete

layer. The complexity of the site forced the

designers to consider 4 options for construction.

Option 1 was to advance by blasting followed by

shotcreting over rock. Option 2 was similar to option 1 ; however, a thicker shotcrete section was required. Due to the continuous

deformation of the clay deposit at sections of the

tunnels, a third option was considered. This

option required 45 cm shotcrete, 60 cm

intermediate reinforced concrete and finally a 60

cm of unreinforced concrete lining.

Option 4, as shown in Figure 27, was planned for

the worst condition since there was evidence of

soil squeezing and also settling of the tunnels

due to extremely soft soil. In this option, two

pilot bench tunnels with a radius of 2.5 m were

excavated along the sides of each tunnel for

rigidity and to stabilize the tunnel movement.

The pilot bench tunnels were later filled with

concrete. In bridge terms, these tunnels act similar to bridge abutments.

The seismic design for the tunnels was based on

experience and judgement. The EPA of 0.4g was used in design, which corresponds to a

return period of about 500 years.

2.3.2 Field Observation

A very brief visual inspection of the Bolu

tunnels was made on December 2, 1999. The

inspection consisted of a drive through of both

bores from the Elmalik portal to the collapsed

sections, approximately 300 meters from the

portal entrance, with limited time spent at the

face of the collapse and in the structurally

incomplete segment leading to the collapse. The

collapsed locations are shown in Figure 28.

They occurred in a section of tunnel passing

through a clay/weak rock zone where a temporary shotcrete lining system was in place.

475
Figure 25: Bolu Tunnel at Elmalik Portal Figure 26: Bolu Tunnel at Asarsuyu Portal
3.0 SUMMARY/CONCLUSIONS

1 ) The FHWA team inspected the roadway

alignment and the physical condition of the Bolu

Viaducts #1 and #2, and Bolu Bridge #2. Also,

the twin Bolu Tunnels damage was viewed from

the Elmalik entrance westward to the collapse.

2) The Diizce earthquake caused considerable

damage to the superstructure of Viaduct #1 and

to the Bolu Tunnels due to the close proximity

of fault rupture (near fault effect). A surface

fault trace was visible between a segment of

Viaduct #1 piers and evidence of high velocity

impulses was observed from the earthquake

records and at the sites.

3) The stations at Diizce and Bolu, 30 km

apart, recorded PGA of 1 .Og and 0.8g

respectively. It is therefore reasonable to assume

that Bolu Viaduct #1, which was not instrumented, located between these two stations

experienced a PGA of at least 0.8g.

4) Significant damage did occur to the EDU's

and bearings of Bolu Viaduct #1 from surface

rupture displacements and ground shaking which

then caused the superstructure girders to translate on their pier tables, narrowly avoiding

total collapse.

5) The positive connections of many EDU's to the superstructure and pier cap of Bolu Viaduct

# 1 were sheared off.

However, cracking of the concrete was noted at one of the anchorages of the EDU brackets that was attached to the superstructure.

9) Complete closure of both Bolu Tunnel bores was noted about 200 to 300 meters from the

Elmalik portal entrances. The collapses occurred

in a section of tunnel passing through a clay/weak rock zone where a temporary

shotcrete lining system was in place. Severe

shotcrete spalling was noted between sets and at the sets along the tunnel wall.

In summary, the close proximity of the fault rupture to the Bolu structures caused significant

superstructure movements relative to the

substructure which resulted in severe damage to bearings and EDU's, in particular, on Bolu

Viaduct #1. However, despite an earthquake

larger than design for, and a fault surface rupture

under the viaduct, collapse did not occur and the

viaduct performed satisfactorily.

Also due to types of damage inflicted on the

bridges by the Diizce earthquake, much can be

learned by the earthquake engineering and

seismological communities. The near fault

effect, the behavior of the EDU's and their

positive connection to super-and sub-structure

are just a few of the factors that can help

engineers in designing and retrofitting bridges to

resist future NAFZ type seismic activity.

6) Shear keys or blocks used to control transverse girder movement of Bolu Viaduct #1,

although severely damaged, appear to have

functioned as intended.

7) Moderate structural damage occurred to the

concrete box girders of Bridge #2. The west- end of the eastbound span translated

approximately 3-1/2 inches to the north relative

to the west abutment.

8) Although the bearing pads on Bridge #2 had

translated about 6-1/2 inches to the north and

about 3-3/4 inches to the west, there was no

significant damage to the EDU's and bearings.

477
j

i

\
Building Damages in Taiwan Chi-Clii earthquake

by

Masaomi TESHIGAWARA", Hiroshi FUKUYAMA^', Hiroto KATO'', Mizuo INUKAI''

ABSTRACT

This paper describes Building Damages in Taiwan Chi-Chi earthquake. Two Specialist

Team of Japan went to Taiwan in order to execute of the emergency risk inspection, etc. Here, the following are introduced : Disaster

situation by both teams and content of the technical assistance cooperation and transition of the earthquake resistant design code of

Taiwan. Still, disaster situations, etc. described

in this paper are not always an outline of the whole damage, since it is based on limited

investigation.

Key Words: Taiwan Chi-chi earthquake

Hyogo-ken Nanbu Earthquake

Building damage

Building Risk Inspection

Building Seismic Code

1. Introduction

Taiwan Chi-Chi earthquake took place in early morning, on September 21st, 1999, with

the magnitude 7.7 (U.S. geological survey

institute) and with the magnitude 7.3 (Taiwan

Central Weather Bureau). This epicenter was

near Chichi town, Nantou county in Taiwan. It became a catastrophe of 2,333 deaths, 10,002

injuries and 39 missing persons, and 9,909

complete collapse buildings, 7,575 middle

damage buildings (Ministry of the Interior,

National Fire Administration, Oct. 13. 2000).

Authors went to Taiwan for 5 days on first of

October from 27th of Septembers, which related

to execute of the emergency risk inspection, etc. while the disaster situation of the building is investigated in this place, as a Japan Disaster

Relief Team specialist team (Team leader : Kazuo Okayama, chief of disaster prevention

section, National Land Agency earthquake) for

the purpose of investigation on the cooperation

needs and recommendation in technical aspect

and administration side. And, another team

entered into Taiwan for 5 days in the 10 days

from sixth of Octobers as Taiwan building risk inspection specialist team (Team leader : Masaomi Teshigawara) formed for the purpose

of the guidance recommendation based on

experience in Japan in the second risk evaluation

of the building which the Taiwan authorities executed in the damaged area. Here, the following are introduced : Disaster situation by

both teams and content of the technical assistance cooperation and transition of the earthquake resistant design code of Taiwan. Still, disaster situations, etc. described in this paper

are not always an outline of the whole damage,

since it is based on limited investigation.

2. Earthquake

(1) Epicenter and scale Generation : At 1:47 a.m. September 21st, 1999.

The scale : Taiwan Central Weather Bureau announcement: magnitude 7.3 U.S. Geological Survey announcement:

magnitude 7.7

1) Head, Structural Division, Department of

Structural Engineering, Building Research

Institute, Ministry of Construction, Tsukubashi, Ibaraki-ken, 305-0802, Japan)

2) Head, Large Scale Structure Testing

Division, Department of Production

Engineering, ditto. 3) Chief Researcher, Large Scale Structure Testing Division, Department of Production

Engineering, ditto. 4) Chief Researcher, Building Engineering

Division, International Institute of Seismology and Earthquake Engineering,

ditto.

479
(the moment magnitude are about 10 times more than Hyogo-ken Nanbu Earthquake's)

Epicenter ; Near Chi-Chi town, Nantou

County, 23.85 N latitude and 120.78 East longitude degrees, 1.1 Km depth (by the Taiwan

Central Weather Bureau).

It is said that the Tamoupu-Hsuangtung fault and the Chelungpu fault which pass through the epicenter moved. Anyway they are almost the

east rised fault of the north-south direction. Duration time : About 40 seconds

Maximum acceleration : 984gal (observation

point of 10.2 kilometer from the epicenter)

Fault fracture surface : 80 kilo X 40

kilometer.

(2) Seismic intensity distribution

Main shock:

Seismic intensity 6(>250gal):

Nantou county Mingchian Village,

Taichung city

Seismic intensity 5(> 80gal):

Hsinchu county Chupei, Tainan county

Yungkang, Chiayi city, Ilan city

Seismic intensity 4(> 25gal):

Pingtung county Chiuju, Taitung county

Chengkung, Penghu county Makung,

Taipei city, Kaohsiung city, Taitung city

Seismic intensity 3(>2.5gal):

Miaoli county Sani, Hualian city

(using 7 class seismic intensity scale in Taiwan which removed seismic intensity of the

7"" class from the old 8 class seismic intensity scale of Japan).

The largest aftershock: 7:52 a.m. September

26th, 1999 generation, magnitudes of 6.8. Seismic intensity of 6: Nantou county Jihyuetan

Seismic intensity of 5: Chiayi county Alishan, Tainan county

Chiali, Chiayi city

(the seismic intensity of 2 in Taipei City).

The large earthquake in a past : Miaoli earthquake (in Hsinchu, Taichung) on

April 21st, 1935. M 7.1. 3,276 deaths.

(3) Earthquake strong-motion record

In Taiwan, there are some earthquake strongmotion record networks which are Central Weather Bureau Strong-motion Network

(CWBSN) with 75 observation points, Institute of Earth Sciences, Academia Sinica (lESAS)

with 109 observation points. Central Weather

Bureau Earthquake Center (TSMIP) with 600

observation points, etc., and many strongmotion earthquake records have been obtained.

Main maximum accelerations of the earthquake

annoncement system by the Central Weather

Bureau are as follows. With the observation, the maximum acceleration rapidly lowers, as it separates from the epicenter.

Nantou ( 10.2 km from epicenter) 973gai

Kukeng( 28.7 " ) 394 '' Alishan( 27.7 " ) 231"

Taichung( 34.6 "

) 222 " Chiayi( 53.3 " ) 166 " Jihyuetan( 12.7 " ) 165 " Hsinchu( 107 "

) 94 " Kaohsiung( 150 " ) 16 " The distribution of strong-motion earthquake

record and damaged area is shown in Fig. 1. According to the spectrum of the acceleration and the pseudo velocity by the strong-motion

records, a bigger acceleration than Kobe is observed in the short period. And, in another

record point, a big velocity is observed in the long period.

3. Damage of buildings

3.1 Outline of damage

Reinforced concrete structures and framed

masonry structures are typical type of the buildings in Taiwan. At others are steel structure,

steel and reinforced concrete structure, brick

structure, wooden structure, soil brick (the dried brick) structure, etc.. It seems that recently

reinforced concrete building are almost used. The damaged building in this earthquake

spreads over 200 km. The features does not seem that the building is suffering from heavy

damage but the buildings of heavy damage or

480
collapse are scattered in the buildings of the

light damage (Photo 1, 2). However, all buildings on the large fault (photo 3) which

exposes from Fengyuan City to Shihkang

Village in Taichung County, at Takeng district in the north-east of Taichung City, from Wufeng

Village, Taichung County to Tsaotun Town,

Nantou County, etc are getting collapsed or

inclined damage (photo 4~6). And, we can say

that it is also a feature that collapse of high-rise

residential building (photo 1, 14) which exceeds

30m high are seen not a few.

In Taichug City which population is about

1,000,000, the exposed fault is 10 km from the center in the city(photo 3). In spite of it, the damage is considerably less than Hyogo-ken

Nanbu Earthquake, and infrastructures, etc. almost normally work. The number of the heavy

and middle damaged buildings in Taichung City

are 496 and 516 buildings, respectively (at present October 13th).

The damage in Taichung City, Taichung

County and Nantou County are as follows.

(1) Taichung City

In south of the city, the damage has been

limited to the houses in which reclaimed land

(photo 14, 15), which has some structural disadvantage at the cross of roads, and the houses in the north near which stands on the

fault (photo 3).

(2) Taichung County

In Fengyuan City, Taichung County which was

located in north about 30km from Taichung City,

Taichung County, there were the fault which

reaches less than 5m heigh observed. And, in Tungshih Town (photo 26), a collapse of smallscale houses such as the one-story house was

mainly observed, removal works of the collapsed houses were advanced including the most intense district of the damage, and the approach by automobiles was not possible.

(3) Nantou County

We inspected in Mingchian Village (photo 27),

Chichi Town (photo 28 and photo 29) and

Chungliao Village (photo 30) of west in Nantou

County and Nantou City (photo 31). Especially, the failure situation of Chungliao Village is serious. There is some zones surrounded by

roads in which almost buildings are collpased

which means the extent of the damage. There was the long space between the reinforcement

bars from the situation of the failure structure. And the buckling of columns and integration of

the oil tin to the under side of beams, etc. were

observed as well as in Taichung City. 3.2 Damage outline of the RC building

(1) Damage of the low-rise RC building

There are many cases which the first floor are used for stores, and over the second floor for houses on low-rise building about 2~5 floor (photo 7). Though generally they are the frame

structures, the walls are often established in 3 planes except for the part which faced the road.

It is the structure with many quantities of the wall which can effectively resist for the earthquake. On the side which faces the road, there are one thing of the 1 span and other thing

of the row house type of the multiple-span. It seems to often establish the boundary walls over

all layers in the multiple-span houses. Generally

the damage of the buildings of this type is comparatively light in photo 8,9. Especially,

though the building in photo 9 was under

construction near the fault, it was non-damaged.

However, large damage such as a collapse was

also observed in the buildings without the wall

at the corner of the road and the buildings

without 2 planes or 3 planes on the first floor.

Still, large damage seems to have been

generated in the building of framed masonry

structure (the structure which established the wall of the brick in the RC frame) which appears

the apparently similar structure for the mortar

finishing.

In the building in Taiwan, the 1 span is often spent as a public sidewalk, and in this case, over

the second floor, it is supported by cantilever beams and columns on the first floor. Therefore,

it becomes the depth direction of 1 span of first floor with (over the second floor some) the type

481
that the wall is not established. In the building

with small spans of the depth direction in this investigation, the damage (photo 11) in which

this seemed to be a cause was not many, and it seemed to be also observed damage such as a collapse, since it becomes a structural type equal

to pilotis.

(2) Damage of the middle-rise RC building

Middle-rise building with about 7,8 floor is mainly seen in the city area. In the building

(photo 12) in which the wall of first floor becomes little the pilotis type for houses, in addition to this, for the over second floor in the building in which the wall is in the corner of

road and store, on the first floor there was the damage in which the first floor collapses, and

the many buildings of non-damage also exists.

(3) Damage of the high-rise RC building

The high-rise buildings which exceeds the 10th floor is mainly seen in multiple dwelling

house in urban area. It is also said that the lower

layer floor is used as a commercial building.

Some structures are the frame structure which

consists of columns and beams, and others are

the frame with structural walls which consists of

bricks in all frame (photo 13). Damage mainly

observed in these buildings is collapse and

heavy damage of most lower layer which

originates from the destruction of the column of

most lower layer which becomes the pilotis type

(photo 14~16) and collapse and inclination damage of the building which originates for it. And, the shear failure and the buckling failure of

columns at the large setback in the middle floor (photo 17), the heavy damage of non-structural members by the excessive deformation, etc. are also observed. There is not always and greatly

the damage of low-rise buildings around these

high-rise buildings. Still, the high-rise buildings

without structural damage also reaches in great

numbers.

(4) Comparison of RC building damages

between this earthquake and Hyogo-ken

Nanbu earthquake

According to the Table 1, the damage of the RC building observed in Hyogo-ken Nanbu

Earthquake'" are also observed in this Chi-chi

Earthquake. For example, photo 18 shows a collapse of the soft storey (pilotis type) of a elementary school, simultaneously, the shear

failure of short columns. Except of these damage features in present earthquake, the following damages are mentioned : Collapse and

inclination of the high-rise residential buildings

and other buildings right on the fault.

Table 1 Comparison of damage features of RC buildings between Hyogo-ken Nanbu

Earthquake and Taiwan Chi-chi Earthquake

Damage features of RC buildings in Hyogo-ken Nanbu Earthquake"*

Taiwan Chi-chi

Earthquake

1) Failure of pilotis ( storey collapse or heavy damage ) many cases

2) Collapse or heavy damage of first floor, not pilotis ( the torsional destroy by

the irregular arrangement of walls and the shortage of shear capacity, etc.) many cases

3) Damage of middle storey according to the vertical direction discontinuity of

each storey's strength or rigidity

confirmed

4) Shear failure of short columns attached handrail walls and hanging walls confirmed

5) Severe damage of columns by shortage of hoops and anchoring shortage of

hoop end.

confirmed

6) Destroy of connections between main bars confirmed

7) Collision between buildings confirmed

8) Destroy of beam-column joints not confirmed

9) Destroy at the column sections of main bar arrangement discontinuity not confirmed

10) Fall down of precast concrete roofs not confirmed

ll)Column yield of frames designed for beam yield many cases

482
(5) Damages of other building structures As for frame structure with infilled masonry

walls, brick structure (Photo 19), wooden

structure (Photo 20), sundried brick structure (Photo 21), the ratio of damage buildings to non- damage buildings were more than the RC

building, and these complete collapse buildings were not a few found. And the collapse of the

light steel building was also observed.

Still, the damage of the steel reinforced concrete structure building has not been

confirmed in this investigation.

(6) Damage of civil engineering structures The severe damage is also observed in civil engineering structure. Destroy of a dam (Photo

22), collapse of bridge (Photo 23) and destroy of

roads by landslides (photo 24), etc. took place.

However, important facilities such as the expressway seem to be designed much strong

which prepared for the emergency, and heavy

damage in this earthquake is not reported.

4. Building seismic design code of Taiwan

Before 1973, the seismic design code of

Taiwan, which referred to seismic intensity

method of Japan was adopted, and it was

allowable stress design for 0.1 design seismic

coefficient. After the war, since the values of

allowable stress became twice, without the design seismic intensity changing, the base shear

coefficient in that time for the ultimate strength

of the building is about 0.15. This means that the base shear coefficient is about 1/2 of RC

building frame structure of the current standard

in Japan.

In 1997, referring to the U.S.A. UBC in 1974,

the base shear coefficient method was adopted.

It is the allowable stress design method until the

revision 1997 in the strength side. By this

revision, the base shear coefficient for the design of usual RC building frame structure in the largest zone factor, was 0.084, the ultimate

strength coefficient became about 0.13, and it became about 83% of its previous standard. The base shear coefficient for the design of

usual RC building frame structure was revised in 1982. And this coefficient became 0.10 in the

largest zone factor including Taichung County,

and the ultimate strength was about 0.15. These

coefficients were equivalent to ones of the code

before 1973. (However, they are this 0.8 times

in Nantou County.)

In 1997, the evaluation of the deformation of

the building was revised, considering the ground

property. The standard base shear coefficient is 0.157, while the ultimate strength design method

is adopted to the RC buildings, and load factor for it is 1.403. That is to say, standard base shear

coefficient of the ultimate strength became

0.157 X 1.403=0.22. Still, as for the steel structure, the allowable unit stress design is adopted, and the design base shear coefficient

for this design is 0.196. After this revision, in the region of highest seismicity, the ultimate

strength base shear of usual RC building frame

structure became about 73% of the base shear in the region where the zone factor is 1 at the current standard of Japan. In Nantou County and

Taichung County which are close to this epicenter, since the region horizontal acceleration coefficient is 70% of the maximum

value, the required strength is about 50% of

Japan. In this revision 1997, the horizontal force

in highest seismicity became 1.5 times of the previous one, while the period of the building was less than 0.3 seconds as mentioned above.

And the horizontal force became equivalent

while the period was more than 1 second, or it was smaller than the previous value depending on the ground classification. In comparison with

the standard of Japan, the 70% strength is required in the short period region. And the required strength is about half in the period from

1 second to 2 seconds, in comparison with the type II ground of Japan. Fig. 2 show the

relationship between the base shear coefficient and period of the RC building with the biggest

capacity by the seismic code after 1982

(Ds=0.3) of Japan and Taiwan.

483
5. Technical assistance and cooperation

(1) Building emergency risk inspection of

Taiwan

The emergency risk inspection shall be

carried out by the two phases. At the r' phase, in seven field (collapse,

inclination, members such as columns and

beams, something falling down, effect to the adjacency building, foundation, others (situation

of the infrastructure)), each damage is inspected

in three ranks (light, middle and heavy). If there

is one field of "heavy" or "middle" rank, the mentioned building is inspected as the rank "no

entry" or "limited entry" respectively, and the red paper or yellow paper are set respectively. In other case, the green paper is set on it as the rank "safety".

At the 2"'' phase, according to the structural method (reinforced concrete, steel, brick, wood),

each evaluation items are established, and each

items are inspected into three ranks (light,

middle and heavy). This inspection method is similar to the emergency damage inspection in Japan.

In Taichung County and Taichung City, the

first phase of emergency risk inspection was

already done or finishing another several days in 29th of September.

(2) 2°'' phase of building emergency risk

inspection

After the T' phase of building emergency risk inspection, there are some problems which are the clarifications of the inspection technique and

the enough guidance about for the criterion to the inspectors, etc. in order to promote the second phase. There was a request from the Taiwan authorities about the inspection guidance

for 3 days by seminar type which is concerned

with the second phase for registered architects who carries out the emergency risk inspection,

etc. in Taichung County and Nantou County.

The second phase is correspondent to the emergency damage inspection in Japan. The

specialist team of Japan was operated for the respond to this request.

By the decision of the high level in the Taiwan authorities on fourth of October, the

priority was given to "investigation for the financial support", not the second phase. So, the second phase was not yet prepared, but the

specialist team was operating for the initial purpose of guidance and recommendation from

the lessons of Japan about the implementation

method of the second phase in damaged area for Taiwan authorities. Problems, guidance and recommendations on

the second phase are as follows.

1) Link with flnancial support

Taiwan authorities had a problem that in a certain local government, the result of the emergency risk inspection was linked with

financial support for damage or apartment house

payment by Taiwan government. Therefore, some people requested the inspection result "no

entry" to inspectors.

The specialist team explained that in Japan,

the main purpose of the inspection was the prevention of the secondary disaster by

aftershocks, and was not linked with the support

to the cost for the reconstruction of housing or, priority occupancy measures to temporary

dwellings, etc.

2) Responsibility of structural engineers

The complaints to the responsibility of the

structural engineers of the buildings inspected "no entry" was strong.

Explanation about this problem is that the reason of damage should be discussed. For example, the building code should be revised when the higher external force than expected was happened, or some damage should depend on the bad construction works. These means that

it is not reasonable that structural engineers

should be responsible with the damage. The

correct reason of damage should be resolved.

3) Removal of buildings inspected "no entry"

What is the dealing with the case when the owner of the building inspected as "no entry"

does not cooperate in its removal ? The explanation to it is that though there are

484
no obligation such as the removal even in Japan on the result of the emergency damage

inspection, referring to the Building Standards

Law, the order of the removal is possible, when

this building is considered to be dangerous.

(3) Outline of questions in discussion in Taipei City and seminars

1) Validity of the level of seismicity evaluation

index Iso. Seismicity input, region

coefficients, etc.

2) Dealing with the building suffered from the damage in the earthquake.

3) Dealing with the building which is not

satisfied with the standard.

4) Role of structural engineers or government for

retrofitting

5) Relation between minimum strength

regulations of the standard and observed

ground acceleration, and relation of the collapse buildings,. Possibility of the authorization of the disaster area, etc.

6) Regulation to decide the removal, repairing or

retrofitting about inclined buildings.

(4) Suggestions from the Japan side.

1) The clarification of the meaning of the inspection system.

*The distinction of emergency risk inspection and financial problems such as

the support.

2) The improvement in accuracy and reliability

of inspection.

•Training for inspection.

•Utilization of the lessons from Hyogo-ken

Nanbu Earthquake disaster. 3) The utilization of the lessons from damaged

area. *Respcts of quality control and construction

rule. •Promotion of structural design and

construction quality control by structure engineers and structural designers.

4) Accurate retrofit and repair

•Enough explanation aobut the distinction of

dangerous building and removal building.

•Utilization of epair and of the retrofit

manual of Japan.

•Utilization of evaluation of seismic

capacity for effect of retrofitting.

6. Conclusion

Taiwan side concretely resolve the problems

which are notification of the meaning of the inspection system, improvement of accuracy and

reliability of the risk inspection, and utilization of the lessons from the damaged area and

accurate retrofitting. These activities needs the knowledge and experience of Japan.

So it is desired that the cooperation of science

and research collaboration and local government

or private cooperation be closely promoted

between Japan and Taiwan.

Acknowledgment

The effort of Hideaki YOKOI, Head, Applied

Seismology Division, International Institute of Seismology and Earthquake Engineering,

Building Research Institute, Ministry of

Constructionx Japan is greatly appreciated, as for the earthquake dimension of this paper,

[Reference]

(1) "1995 Hyogo-ken Nanbu Earthquake

damage investigation report (outline

edition)". Building Research Institute, Ministry of Construction, March 1996.

485
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•

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• Tungshih

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Taichung City

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• Wufcng

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Taichung Counly

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tFaull and geological lypc

riKKXIK MioUl

Fauki ObMrvcd byh on th< trownd

( Are with underline : Investigated area)

Fig.1 Damaged are and maximum accelerations observed

(Ref. 'NIKKEI Architecture', 18 Oct 1999, pp. 107-122 in Japanese)

Fig.2 Comparison of the design base shear coefficient after 1982 in Taiwan with

the vibration coefficient (at Ds=0.3) in Japan.

486
Photo 1 Collapsed high-rise multiple dwelling

house (Tungshih, Taichung County).

The 12-floor RC building collapsed by the destruction of the pilotis column. The damage of

the low-rise building of the vicinity is very

greatly nonexistent.

Photo 3 Fault (Takeng, Taichung City)

The displacement of 3~5m vertically and about 4~8m horizontally has been generated

Photo 5 Building just above fault (Fengyuan

City, Taichung County)

The ground behind building is rised. The building

in this vicinity got same damage

Photo 2 Situation of damage from the sky (Puli, Nantou County)

Though it is comparatively a region of heavy

damage, it is the situation that the building thai almost buildings are light damage and collapsed

is scattered .

Photo 4 Building collapsed just above fault (Fengyuan City, Taichung County)

Building of brick is perfectly destroyed

Photo 6 Building just above fault (Shihkang,

Taichung County)

RC multiple dwelling house of the new

construction inclined by the fault. The pile is not used. The damage of the superstructure is very

greatly nonexistent

487
Photo 7 Typical street in Taiwan (Tungshih,

Taichung County)

Store on first floor and houses over second floor are used. The 1 span in the roadside of first floor

is used as a public road.

Photo 9 The low-rise building (Fengyuan City,

Taichung County)

It is under construction near the fauh of photo 4,5. The structure is almost non-damage

Photo 1 1 Damage of the column in the first floor setback division (Tungshih, Taichung County)

Photo 8 Typical low-rise building (Fengyuan

City, Taichun County)

In one direction of the building, RC wall except

for the 1 span in the roadside of first floor without

the window has been established. Non-damage.

Photo 10 The low-rise building (Fengyuan City,

Taichung County)

The wall was distributed irregularly even in the low rise, and there was heavy damage in the building with small walls on first floor.

Photo 12 Middle-rise building which is built in the corner lot where first floor collapsed

(Taichung City)

By there being small twists by the localization of

the wall and wall quantities of first floor, it is supposed destroying the column of first floor. Large damage is not observed in the building in the vicinity.

488
Photo 13 Damage of brick wall of the high-rise

residential building (Fengyuan City, Taichung

County)

Large response displacement occurs, since there

is no RC structural wall, and it is supposed the heavy damage of the brick wall.

I Photo 15 High-rise multiple dwelling house in which first floor collapsed (Taichung City)

[ Destruction situation of the first floor column in photo 14. The damage of the lower part is being

concentrated at first floor.

Photo 17 Damage of the middle floor of the high-rise multiple dwelling house (Fengyuan City,

Taichung County)

The damage is concentrated in the column of the middle floor of the setback largely.

Photo 14 High-rise multiple dwelling house in which first floor collapsed ( Taichun City)

Photo 16 Damage of the first floor column of

the high-rise multiple dwelling house (Tunshih,

Taichung County)

15-floor building of design and construction equal

to photo 1. A collapse was avoided, and the column of first floor has damaged.

Photo 18 Elementary school in which first floor collapsed (Tunshih, Taichung County)

The wall of first floor was pilotis type, and the nonstructural wall was also attached to the column, and it became short column.

489
Photo 19 Damage of brick structure (Shihkang,

Taichung County)

A collapse was avoided, and large crack was

generated in the brick wall. The brick fence of

the front has collapsed.

Photo 20 Damage of the wooden building

(Tungshih City, Taichung County)

The wooden building of tiled roof is observed. Though the building of the photo remained for an

inclination, it was also seen the result to collapse

in the vicinity.

Photo 21 Damage of sundried brick structure (Tungshih, Taichung County)

It is completely destroyed. It is near the building of photo 7.

Photo 22 Destruction of a dam (Shihkang,

Taichung County)

The fault crossed the dam.

Photo 23 Collapse of bridge (Shihkang,

Taichung County)

It is an upstream of photo 22, and collapse is considered to be the effect of the fault. The

damage of the bridge was observed not a few. Photo 24 Landslide in the mountain ( near

Shuili, Nantou County)

Landslides are being generated in everywhere,

and the road has been blocked.

490
Photo 25 Fault (suburb of Fengyuan City, Taichung county)

The part of the bridge rises and moves to the horizontal direction, too. ( The overpass rail which

is shown by the red broken line was straight originally.)

The right side at the building which was built across the 9 m road rises and moves to the side of

the building on the left side. The road between buildings disappear and the car in parking is put in.

Photo 26 Fault (suburb of Fengyuan City, Taichung county)

The side of the left back at the farm field which was to be plain rises. The dug rice dies and being discoloring to the brown of it. The damage on the side (the east side) where the fault rose is big whereas it doesn't seem to suffer heavy damage on the west side. 491
Photo 27 Collapse of high-rise residential building (Mingchian Village, Nantou County)

One of the 12nd floor building RC building falls down while another which is the same

as the building in the back is damaged lightly. During removal work

Photo 28 Collapse of high-rise residential building (Chichi Town, Nantou County)

The top and bottom of plain columns on the 1st floor of the 3rd floor RC residential building floor destroys in bending failure, and it is swaying to the right.

Photo 29 Building damage under construction (Chich Twon, Nantou County)

RC building under construction is finishing the Isl, and 2nd floor of casting concrete and the 3rd

floor is casting.
Photo 30 Area where a lot of buildings suffered damage above the heavy (Chungliao

Village, Nantou County)

The building which faced a street suffers completely heavy damage.

The oil can in the concrete which was made a topic by the newspaper.

It is explained by the related engineers when the oil can is the one to be using for nonstructural parts such as the beam, in order to make wipe and isn't using for the main

structure body.

Photo 31 Damage of high-rise multiple building (Nantou City, Nantou County)

About 50 % of the columns which faced a road in the side at the 12nd floor RC

buildings which are built in the corner place is being heavy damaged.

493
Photo 32 Damage of high-rise multiple building (Fengyuan City, Taichung County)

Damage of the three 22nd floor RC residential buildings (the lower floors are commercial

facilities) which are built as the surrounding of a court. Damage of the non- structural walls on the middle floors and damage of the column tops on the 1st floor.

Photo 33 Damage of high-rise residential building (Fengyuan City, Taichung County)

One of the 12nd floor RC residential building which are the same as the right one in this photo

falls down in the 1st floor. Removal work and column.

494
DAMAGE OF EARTH RETAINING STRUCTURES

IN THE 1 999 CHICHI, TAIWAN EARTHQUAKE

Ken-ichi TOKIDA'\ Osamu MATSUO'\ Shinya NAKAMURA^'

ABSTRACT

This paper reports damage to earth retaining

structures due to the 1999 Chichi Taiwan earthquake,

based on the post-earthquake field survey. Types of

retaining stnictures include conventional concrete

retaining walls, geosynthetics-reinforced soil retaining

walls and segmental retaining walls. Details of the damage are described and their feilure mechanism and

implications are discussed.

KEY WORDS: earthquake,

retaining wall,

damage

l.JNTHODUCTION

At 1:47 ajn. on September 21st, 1999, a strong

earthquake occurred with its epicaita- in the Nan-Tou

prefecture in central Taiwaa Magnitude ofthe earthquake

is M]=73 (Taiwan coitral meteorological oflBce) or Ms=7.7 (U.S.G.S.). The earthquake caused feult with a

total laigth of 80 kilomdas. It also caused loss of mere

than 2,400 people's lives and damage to more than 20,000

houses, e^densive geo4ia2ards and damage to various OTgine««i stmctures. The authcas conducted a field

survey z&er the earthquake focusing on damage to soil

stmctures.'^ The survey was conducted as a part of the

cooperative research project, 'Ttevelopment of

Earthquake and Tsunami Disaster Mitigation

Technologies and Their Integration for the AsiaPacific Region", budgeted by Science and

Technology Agency of Japan. ITiis papa txiefly

desoibe the damage to several types of retaining walls

ccmstructed along roads and tiieir feiline medianisnis are

discussed The location of the three sites surveyed in this paper is shown in Figure 1.

2. CONVENTIONAL CONCRETE RETAINING

WALLS

Among the types of conventional concrete

retaining walls, the leaning- and gravity-type retaining walls seem to be most popularly constructed in Taiwan.

A small number of cantilever-type concrete retaining walls and geosyntiietics-reinforced soil retaining walls were found within the authors' field survey. The

cantilever-type concrete retaining walls were observed

in flat areas, not in mountainous/hilly area. This might

be due to the fact that this type of retaining walls

requires a considerable width of base plate and when

constructed in steep slope side the need for setting-back

slope cutting would increase the construction cost For

the same reason, the leaning- and gravity-type retaining walls are generally preferred in mountainous/hilly area. Within the survey, no damage to the cantilever-type

retaining walls was found. According to Dr. Chou, a

local consultant engineer, this type of walls has been

usually designed seismically in Taiwan.

In contrast, serious damage to a number of

leaning- and gravity-type retaining walls was found in

hilly areas. Most ofthe damaged walls were located on

steep slopes and their typical failure modes were sliding

and tilting, occasionally associated with collapse at

1) Director, Earthquake Disaster Prevention Research

Center, Public Woiks Research Institute, Ministry

ofConstruction, Tsukuba, Ibaraki-ken, Jq)an 2) Head, Soil Dynamics Division, ditto 3 ) Senior Research Engineer, ditto.

495
construction joint

Photograph 1 shows the typical damage of the

gravity-type retaining wall. The site is located midway

of a mountain road connecting Tsaulin and Chushan

(Site© shown in Fig.l). As shown in Figure 2, the road is a cut-and-fill constructed on a steep slope with a slope angle of 50 degrees to the horizontal. The

retaining wall with a height of about 7 meters was

standing ahnost vertically, thus may be categorized as

gravity-type retaining wall. However, they seemed too

slender. The base of the walls were insufficiently

embedded into the foundation soil and no special

treatment to strengthen the bearing capacity was made.

It is therefore presumed that the walls had only very

small safety margin against external disturbances. Two longitudinal segments of the retaining wall

completely tilted down the steep slope and the neighboring segments were distorted as shown in

Figure 2. These remaining walls appeared to have

tilted back, implying that the base sUde movement

surpassed the tilt-forward movement.

The downslope foundation soil was loose and

trunk of a tree located about 5 meters down from the

retaining wall was bent at a height of 1 meter from tiie ground surface, while the upper part of the trunk

standing vertically. In addition, asphalt pavement was

overlaid in several layers with a total thickness of 50 cm. These evidences convince us that there had been

landslides associated with limited amount of

movement during the past heavy rainfalls, which was

confirmed by local people as well. The observations suggest us that the factors

leading to collqjse of this gravity-type retaining wall would have been:

(a) Foundation soil was potentially unstable to have

exhibited landslides during past rainfalls.

(b) No special treatment had been made to strengthen

the foundation against bearing failure.

3. GEOSYNTHETICS-REINFORCED SOIL

RETAINING WALLS

Photograph 2 shows the cut slope covered by

geosynthetics- reinforced soil retaining wall collapsed

during the earthquake. The site is located near the

fixDnt gate of Chi-Nan International University in PuLi (Sited) shown in Fig.l). As Ulustrated in Figure 3, the slope with a total height of 80 meters had been cut

in a 60-degree slope and was refilled witli the geosynthetics-reinforced soil retaining wall to cover

the cut slope.

In the damaged zone, the surface was retained with the geosynthetics-reinforced soil retaining walls. The reinforcement, geogrid, was placed in every 80 to 120 centimeters layer, while wrapping around the

stack of soil bags at the slope face. The retaining wall was composed of six or seven sub slopes with a height

of 10 meters each and horizontal steps of two to three meters wide. The reinforcement length ranged fixjm 4

to 13m as in Fig.3, while the wrap-around length was

1.5 to 1.8 meters, according to the authors' survey. It should be noted that according to local people the upper portion ofthe wi^nal slope feiled along an inclined weak layer due to a rainfell during flie cut and fill wcsks. The failure mode looked like the following:

the upper five sub-slopes slid down the slope with

their configuration maintained, while the bottom one

or two sub slopes were hidden behind die overlying

sub slopes. At lower portion, wrap-around

reinforcements were pulled out and soil bags and

interlayer backfill soil spilled out

Although no detailed information is available

at tills moment on the failed slope, the authors assume

a possible failure mechanism as follows (see Figure

4): Figures 4(a) and (b) show the forces and sti^esses

acting on the soil retaining wall before and during the

earthquake. The resultant force acting on the bottom

plane oftiie wall before the earthquake (denoted as Qs

in Figure 4(a)) is given as a sum ofthe self-wei^t and

496
tiie earth pressure at the interface with the cut slope.

During the earthquake, horizontal inertial force acted on Ae soil retaining wall, thereby the normal stress and shear stress at the interface reduced. These

changes would have caused an increase in the vertical component of the resultant force acting on the bottom

plane (denoted as Qe in Figure 4(b)), an increase in

the horizontal component of Qe, an increase in the

inclination of load Qe, and shifting-forward of the

load eccentricity. This load, Qe, would have caused

bearing capacity failure or excessive deformation of

the bottom part of the soil retaining wall rather than

the foundation soil (the foundation at this site seemed

very stiff). Once the bottom part collapsed or deformed, the downslope deformation would have

progressed to tiie overlying part. The observations sugges^c us that bearing c^^acity at

the base of the reinforced soil retaining wall is very

important, and suflBcieait length of reinforcement

OTibedment and wrapping-around are necessary. This

type of seismic instability is likely to occur in

geosynthetics-reinforced soil retaining walls with

short bottom length relative to the height

4. SEGMENTAL RETAINING WALL

In the middle of the prefectural road No. 129

connecting Taichung and Dungshr, there was another type

of the reinforced soil retaining wall, Le., segmental

retaining wall which were damaged as shown in PhotDgr^h 4 (Sited) shown in Fig. 1).

This is a type ofretaining wall in whidi bk)cks are

piled up in the wall surfece, FRP (fiber reinforced plastic)

ccKinecting rods are inserted between each block to

COTStrain the horizontal movCTiCTit and geogrids are fixed by the rods. In recent years, it is widely used in NcHlh

America and Europe, while rarefy used in J^pan.

Figure 5 shows the oxjss section of the distorted

walL In this site, the reinfcacements (geotextile) wae

placed with a vertical intoval of 80 an, and four pieces of

ccxicrete block with dimensions of 20 an hi^ 50 an

wide and 30 cm deep (see Photogr^ 5) were stacked at

the surface. The figure also shows that the stack of

concrete blocks buckled out at a mid4iei^t

Disconnection of the reinfacement fiom the COTmecting

pins and subsequent extrusion of backfill soil woe also observed (see Photograph 6). Crudied stc«ie was used

bdiind the facing, presumably for drainage axisidaation.

Attaition ^ould be paid that bulging of the

caicrete block feeing are located at the mid4iei^ ofthe

reinfcMTcement interval. The buckling should have been

induced by the earth pressure during the earthquake. Ifwe

assume that the feeing behaved like a ri^d wall, the bending moment due to earth pnessure takes maximum at

the mid-hei^ As ti^ reinfOTcement interval becomes

laige, the resistance of the block feeing against buckling

becomes less. Althou^ the design {xocedure for the

feeing has not yet beoi developed, it is conventiaially

recommended in NcHth America to place reinforcements

in evay three blocks (Leschchinski, 1 999 ^.

This damaged case suggests us that the interval of

80 an comesponding to the hei^t offour pieces ofblodc was too large to resist against the earthquake. The aufliors do not know the reason wlty an interval of 80 an was

adopted. If an en^ea- caisidos that the use of

reinforcement with lower straigth associated wifli smallo* pacing can be r^laced by use of reinfOTcement with

larga- straigth associated with larga- spadng, he would

prefer the latter for the reason ofcost and woriiability.

Therefore tiie observations suggest us that some

restrictions sean to be necessary on liie intaval of the

reinfOTConent Also sufiBciait camectiOTi ^rength

betweai each blocks or block and ^©textile are cniciaL

5. SUMMARY AND CONCLUSIONS

Post-eaithquake fiekl survey was ccnducted for

various types of soil r^aining stnictures tiiat woe

damaged by tte 1999 Qiidii eailhquake. The results of

the survey m^be summarized as foUows:

1) Among the convaitional types of conoiete retaining

walls, damage to gravity-type retaining walls woe

497
frequaitly found in mountainous areas. Most of the damaged walls were located at potaitially unstable steep

slopes and they looked sloider in fHtpation to the height

Foundations w^e not treated properiy to strengthen the

bearing capacity.

2) Damage to wrap-around type geosynthedcs-reinforced

soil retaining wall was also observed Possible reasons for

the feilure would be sliding feilure of the CHiginal slope,

bearing feilure at the base of the reinforced soil retaining

wall, and insuflBcient length of reinforcement embedment

and wr^ing-around.

3) In the case ofdamaged ^gmaital retaining walls, large

interval of the reinfcMcements and insuflBcient cainection

strength between the segmental wall and the

reinfcMx:ements seemed to have caused feilure.

Furthermore, the results ofthe field survey suggest us the importance of tfie following considerations in

design and constmction ofretaining walls.

1) When leaning- or gravity-type retaining walls are

ccMistructed in mountainous regjoi, proper evaluation of

bearing capacity in the foundation slope and necessary

treatment are crucial to secure stability not only during

heavy rainMs but also during earthquakes.

2) Earthquake-resistant design prcx:edure for the wall

surfece portion shouW be develqxd firrther in the case of

segmental retaining wall and geosy^thetics^^einfo^ced

walls. And appropiate constructicwn wok is aucial to the

security ofthese walls. The case records will be studied further in the form of back-analysis to verify and improve the

existing seismic design procedures for the types of soil

retaining structures. The construction works of the reinforced soil wall have been increased in Japan as Fig.6. Therefore, a manual titled as "Design and Construction Manual of

Reinforced Soil Using the Geotextile" was issued in February 1993, as a result of cooperative research

project between Public Works Res. InsL and private

firms. Since then, the reinforced retaining walls have

been experienced several earthquakes such as Kushirooki eartiiquake in January 1993 and Han-Shin Aw^i

Earthquake in January 1995 in Japan and demonstrated

hi^ performance^^

The manual was revised in February 2000

based on recent experiences as shown in Appendix

The manual will be used efifectively for fliture constmction ofreinforced soil structures.

REFERENCES

1) O. Matsuo, S. Nakamura: Report of the

field survey on the 1999 Chichi Taiwan

earthquake. Technical memorandum of PWRI,

No.3707, March 2000 (in Japanese)

2) Leschchinski, D. (1999): Private communication to Tatsuoka

3) F.Tatsuoka, M.Tateyama, and J. Koseki:

Performance of soil retaining walls for railway

embankments, A Special Issue of Soils and

Foundation on Geotechnical Aspects of the January 17, 1995 Hyogoken-Nambu Earthquake,

pp.311 -324, 1996.

4) Public Worlcs Research Center. Design and

Constmction Manual of Reinforced Soil Using the

Geotextile, February 2000 (in Japanese)

498
(Fengyuan)

O

(Taichung)

O (Dungshr)

/ (Chelungpu Fault)

S/ (Tsaotun)

(Nantou)

^

1

O

(Puli)

Epicenter

1^

(Chichi)

(Sun-Moon Lake)

^

(Shuill)

0

I (Chushan)

O (Tsaulin)

50km

Fig.l Central part of Taiwan

499
Photo. 1 Collapsed gravity-type concrete retaining wall

(Sitel in Fig.l)

500
Photo. 2 Front view of the collapsed geosynthetics-reinforced soil

retaining wall(Site2 in fig.l)

Photo. 3 Failure of the reinforcement

i

I'

f'

501
(unit: m)

Fig. 3 Cross section of the failed geosynthetics-reinforced soil retaining wall

502
Photo. 4 Collapsed segmental wall(Site3 in fig.l)

0.3m

Fig. 5 Collapsed segmental retaining wall

503
Photo. 5 Shape of the block
aooooocoooooooa>CTia><7iCftaiCT)a^04

Year Fig. 6 A trend of construction works of

reinforced soil wall in Japan

505
Appendixl List of participants of the field survey

Osamu MATSUO: Head, Soil Dynamics Division, Earthquake Disaster Prevention Research Center, Public

Works Research Institute, Ministry of Construction, Tsukuba, Ibaraki-ken, Japan

Shinya NAKAMURA: Senior Research Engineer, ditto. Fumio TATSUOKA: Professor, Department of Civil Engineering, University ofTokyo, Kongo, Tokyo, Japan

Masaru TATEYAMA: Senior Research Engineer, Railway Technical Research Institute, Kokubunji-shi,

Tokyo, Japan

Chin-Chuan HUANG: Associate Professor, National Cheng-Kung University, Tainan, Republic of China

Ichiro FURUTA: OYO Corporation, Nagoya-shi, Aichi-pref, Japan

Appendix2 Abstract of the Design and Construction Manual 2000 for Reinforced Soil Using

the Geotextile, Japan

The abstract of the revised manual especially on the earthquake design procedure is shown as follows.

Classification of Reinforced Soil Structure

In this manual, the geotextile reinforced retaining soil is divided into two types; reinforced embankment and

reinforced soil wall, depending on the slope gradient.

The definitions are given as follows.

(1) A Reinforced Embankment

This applies to the fill with a slope gradient milder than 1 : 0.6. In the design, the stability of reinforcing

area against earth load is not necessary to be considered. Stability check modes are in Fig.l.

(2) A Reinforced Soil Wall

This applies to the fill with a slope gradient steeper than 1 : 0.6. In the design, the stability of reinforced

area is examined on sliding, pull-out and rupture modes, induced by earth load. When the height of the

reinforced soil is less than 8.0m, the seismic stability examination can be omitted, because it can be

functionally safe for the earthquake of the usual scale, conditioned that design and construction are

properly made. When the reinforced soil wall is located on crucial link of highway network and is

considered, if once damaged, to take long time to restore, the seismic stability check should be carried out

Stability check modes are in Fig.2.

Types ofthe Reinforced Soil Wall

Typical types ofthe reinforced soil wall are shown in the manual as wr^ing-around type (Fig3), steel fiame

type (Fig.4), concrete panel type (Fig.5), concrete block type (Fig.6) and other types (Fig.7). Earthquake Resistant Design ofthe Reinforced Soil Wall

Fundamentally, by replacing the seismic load with inertial force, the approach of tfie usual design is ^plied.

The seismic stability check is based on the pseudo-static seismic force method.

Basic seismic design procedure is in Fig.8. And concrete content is as follows.

506
1) Design seismic intensity

kfa: Seismic intensity

l^jo: Standard seismic intensity (in Table 1)

c^: Regional correction coeflScient (1 .0, 0.85, 0.7)

Table! Standard horizontal seismic intensity

Type of Ground I n in

Middle scale earthquake 0.12 0.15 0.18

Large scale earthquake 0.16 0.20 0.24

2) Seismic design safety factor.

Table2 Seismic design safety factor for reinforced soil wall

Types of safety factor Design safety factor Always Earthquake

Internal stability Sliding circle ^12 ^1.0

TensUe force ofgeotextile ^1.0 ^1.0

Pull out of geotextile ^2.0 ^1.2

Extemal stability SUding ^1.5 ^12

Tilting e^L/6 e^I73

Bearing capacity ^2.0 ^1.5

Wholly stability Sliding include ground ^12 ^1.0

Notes) e: eccentric distance from tfie center ofthe bottom ofreinforced soil wall

L: bottom width ofreinforced soil wall.

507
(c) Sliding, outside the reinforced area

(a) Tearing off (b) Pulling out

Fig.1 Stability check modes of reinforced embankment

Slip surface Tear off

(a-1 ) Tearing off

Wall

surfacfe* Connection

Pulling out

(a-2) Pulling out (a-3) Tearing off at connection

(a) Internal stability

(b-2) Tilting

(b) External stability

li '

1 1 *^

lit

lit : _

1 |i " "J^^ Earth

~ V pressure

(b-3) Bearing capacity

(c-1) Sliding include surcharge bank (c-2) Sliding include ground

(c) Wholly Stability

Fig.2 Stability check modes of reinforced soil wall

508
Fig.3 Wrapping-around type

509
Geotextile

Fig.5 Concrete panel type

Geotextile

Concrete

Projection Connection pin Stuffing

ng.6 Concrete block type

510
'Connection

(a) L shape retaining wall type (b) Gabion type

Fig.7 Other types

511
Start

Internal Stability

External stability

Wholly stability

Set up of design seismic load

Calculation of total tensile force

Selection of geotextile type

Examination of geotextile interval

Examination of geotextile length

Stability analysis of wall surface

Stability analysis on geotextile length

against sliding, tilting and bearing capacity

Reconsideration of geotextile length

Check of geotextile arrangement by sliding circle analysis

Change geotextile type and location

Rg.8 Procedure of seismic design of reinforced soil wall

512
HURRICANE MITCH RECONSTRUCTION PARTNERSHIPS AND PROGRESS

William E. Roper'

'Science Advisor for Sensing, Imagery, and Knowledge Systems

National Institute of Standards and Technology

ABSTRACT

A review of reconstruction efforts since

Hurricane Mitch occurred and the application of

information collected during the response stage

are presented. A sample of geospatial

technologies applicable to disaster management

is presented. The application of Geo-spatial

technologies following Hurricane Mitch in Central America is presented. Natural disasters

have a major impact, globally and within the

United States (U.S.) causing injury and loss of

life, as well as economic losses. To better address disaster response needs, new approaches

to leverage technological capabilities are described to improve disaster management.

Geospatial analysis is one of these important

capabilities.

KEY WORDS: Hurricane Mitch; Disaster

management; Central America; Information

Infrastructure; Disaster recovery; mapping; Geospatial library; LIDAR; Geographic Information

Systems; Global Positioning System; satellite sensors; 3-D fly through.

1. INTRODUCTION

This paper will focus on an update of the

reconstruction efforts that have occurred since

the devastation of Hurricane Mitch in the fall of

1998. Particularly, the application of geospatial

technology and methods to assist or support

recovery and reconstruction activities will be

reviewed and assessed. There has been

significant capability growth in the geospatial

areas of remote sensing, spectral analysis, global

positioning systems (GPS), geographic

information systems (GIS), and modeling and

simulation techniques (Roper, 1998). Each adds

important value in characterizing infrastructure,

risk areas, disaster zones, and control points that are essential to rapid deployment of scarce resources in the most effective manner.

Natural disasters are a constant threat to mankind on a global scale. Global disaster costs

are continuing to rise. Annual global economic

costs related to disaster events average $440

billion per year (World Disaster Report, 1996)

with floods being the major cause. In the U.S.,

the number of lives lost due to natural disasters

has been decreasing over the last several

deceased, largely because of advances in disaster

indication and warning capabilities. In terms of

damage to property, however, the trend is reversed. For the period 1992-1996, the average

cost of natural disasters in the U.S. has been

$54.3 billion, with hurricanes and earthquakes

tied as the leading cause. These rising costs are the combined result of increased urbanization,

particularly in high-risk coastal areas, and the

increased complexity and size of our

infrastructure.

The loss of life and property continues to rise in many regions of the world because of these

events. One example is the Bangladesh weather

event of 1970, when a tropical cyclone slammed

into its delta region killing 300,000 people

(Tobin and Montz, 1997). The crop losses were

estimated at $63 million, and more than 280,000

cattle were drown (Burton, Kates, and White,

1993). The rich delta soil is an agricultural

513
resource that still draws people to settle there; therefore a recurrence of this type of weather

event could likely pose a similar disaster.

2. GEOSPATIAL RESEARCH

COMMUNITY INVOLVEMENT

The flow of disaster related information could be

conceptually shown as a seven-step process as shown in Figure 1. The geospatial research

community may have a role in each of these

steps, depending on the particular disaster

situation.

This could include assistance in defining the

problem through current image analysis.

Assistance in determining collection and

analysis methodologies assists in defining the

requirement. A key area of support is providing

supporting data in the form of maps, imagery,

and spectral information. An even more

important support area is data exploitation. This

includes the processing of digital data, image

integration, feature classification and attribution,

classification output, accuracy assessments, and

post-processing operations. It is most

technically challenging step in process. The

decision support phase includes geospatial

visualization, merged data analysis, and

specialized decision support products. This

involves the synthesis of data types in order is generate data layers (such as, soils, vegetation,

terrain) along with models and simulation

techniques that use the various data layers.

It is during this phase that the concept of virtual forums could take place. These forums would

allow expert input from multiple locations to be

jointly focused on problem solving for the

disaster situation. Within the geospatial

community this could include virtual fly-through

support showing the impacts of possible

decisions on the natural environment, and on the

population, including those with special needs.

It also could include analysis of changing

situational information, and image-based change

detection and analysis. In the final two phases,

tailored map products and GIS overlay

information can be used as integral parts of

directives and guidance to disaster managers in the field. These specific products could be maps

delineating evacuation routes, area identification

for damage assessments, or point locations for water distribution.

The geospatial research community has typically

been interdisciplinary, including the natural

sciences, engineering, architecture, land use

planning, photogrametry, etc. However, in the

disaster management arena, broader

interdisciplinary teaming will be required. This

could include public policy and health

specialists at the Federal, state, and local level.

It may include other medical specialists, police,

National Guard, and communications experts.

The greatest potential for loss reduction is during the mitigation phase (Disaster

Information Task Force Report, 1997), when

communities can be made more disaster

resistant. The largest share of costs, however,

are directed toward the recovery phase, where

good mitigation principles also need to be put

into practice rather than just rebuilding to be

impacted by a similar disaster in the fiiture. Geospatial products and location specific tools developed during the response phase of the

disaster may be excellent vehicles for planning

and implementing effective mitigation.

3. HURRICANE MITCH STORM TRACK

AND CHARACTERISTICS

Hurricane Mitch, the strongest October

hurricane ever recorded, formed in the southwest

Caribbean sea from a tropical wave about 360

miles south of Kingston, Jamica, late on October

21. The system initially moved slowly

westward and intensified to a tropical storm

(Atlantic Hurricane Summaries, 1999). Mitch

514
then moved slowly northward. Its course changed to north northwestward on the 23^^ and

24'*' while gradually gaining strength. Early

October 24, Mitch became a hurricane. Later

that day, as it turned towards the west, Mitch

began to intensify rapidly. In about 24 hours, its central pressure dropped 52 MB to 924 MB by

the afternoon of October 25. Further

strengthening took place and the central pressure reached a minimum of 905 MB about 40 miles

southeast of Swan Island on the afternoon of

October 26. This pressure is the fourth lowest

ever recorded in an Atlantic hurricane this century. It is tied with Hurricane Camille in 1969. This is also the lowest pressure ever observed in an October hurricane in the Atlantic

Basin.

At its peak, the maximum winds were estimated

to be 180 mph, a strong Category 5 hurricane.

After passing over Swan Island, Mitch began to gradually weaken on October 27 while moving

slowly west, then southwest toward the Bay

Islands off the coast of Honduras. The center passed very near the Island of Guanaja wreaking

havoc there. From mid-day on the 27"" to early on the 29'^ the minimum central pressure rose 59 MB. The center of the hurricane meandered

near the north coast of Honduras from late on

the 27'*' through the 28"' before making landfall during the morning of the 29"' about 70 miles

east of LaCeiba with 100 mph winds. Mitch

moved southward over Honduras, weakening to a tropical storm early on the 30"'. Mitch moved

slowly over Honduras and Guatemala on

September 30-31, gradually weakening to a depression. The storm generated torrential rains over portions of Honduras and Nicaragua where

the associated floods were devastating. Some

heavy rains also occurred in neighboring

countries.

Although Mitch originally dissipated near the

Guatemala/Southeast Mexico border Sunday

afternoon, November 1, the remnants continued

to produce locally heavy rainfall over portions of

Central America and eastern Mexico for the next couple of days. On November 3, a low-level

circulation became evident in the eastern Bay of

Campeche and an Air Force Reserve

reconnaissance aircraft investigating the system

reported tropical storm-force winds and a 99-

MB central pressure. Mitch had regenerated

into a tropical storm on the afternoon of the

3"^*^ while located about 55 miles west-southwest of

Campeche, Mexico. Mitch weakened to a depression early on the 4"' as it moved inland over the northwest Yucatan Peninsula. The

center re-emerged over the south central Gulf of

Mexico by mid-morning on the 4"' regaining

tropical storm strength. Mitch began to accelerate to the northeast as it became involved

with frontal zone moving through the aster Gulf

of Mexico. Mitch made landfall on the morning

of the 5"' in southwest Florida near Naples with

maximum sustained winds near 60 mph. By

mid-afternoon of the 5"', Mitch moved offshore

of south Florida and became extra tropical.

Figure 2 shows the storm track of Hurricane

Mitch.

4. INITIAL GEOSPATIAL SUPPORT OF

MITCH RECOVERY ACTIVITIES

4.1 Initial Response Efforts

U.S. military forces were some of the first international support units on the ground

conducting recovery operations. As a result, a number of requirements for geospatial products were identified early in the operations

(Operations Summary, January 1999). In support

of these requests, maps, country studies, water

aerial appraisals and other available materials were collected and provided to the U.S. Army

Corps of Engineers (USAGE) Operations Center

for use in immediate recovery operations,

planning any additional deployments and fixture reconstruction efforts. Separate requests for ground water resource data were received from

515
support elements of the Second Mare

Expeditionary Force and two Air Force Red

Horse Squadrons (well drilling units) in Honduras. Maps and water area appraisals were

provided to two Marine Expeditionary Forces. Water detection data was also provided to the

Air Force Unit.

Initial analytical activities focused on assessing

road and bridge damage in Honduras. Satellite imagery was reviewed and road and bridge

assessment data were tabulated and served over

the communication links. These data were

keyed to a regional map, provided by the

National Imagery and Mapping Agency

(NIMA), to which a grid was added as an

orientation aid for the user. Photomaps of the

Soto Cano Air Base and a slide area along the

Chaluteca River in Tegucigalpa were also

prepared and provided to the Army Corps of

Engineering Operations Center. The 30"^ Engineer Battalion, also in Honduras, requested

specific Digital Topographic Data, and loan of

high capacity hard drives to process data. The

Topographic Engineering Center imbedded

current imagery data on the affected areas onto

14 CDs and provided them to the 30'*^ Engineers.

These CDs contained Landsat and Spot imagery

(Collier, 1999).

4.2 Joint Government and Private Sector

Support

A cooperative effort of U.S. Government

agencies and the private sector developed a

Digital Atlas of Central America to assist in recovery efforts and planning for reconstruction.

The geographic information system (GIS) based

product provided a framework for displaying

and organizing information in multiple layers

useful to those responding to the disaster. Products were produced showing, for example,

the path of Hurricane Mitch over a map with

country boundaries and color-coding to illustrate terrain elevation (Digital Atlas, January 1999).

4.3 Geospatial Technology Applications

For the initial reconstruction phase, a number of

geospatial products are needed to enhance

planning and project executio. Virtually all activities associated with the reconstruction

effort required large-scale (high resolution) base

maps showing topography, drainage, and the

location of roads, bridges, and massive sediment

deposits from the flooded Rio Choluteca River

within the city limits is of particular concern

with the fast approaching rainy season. Also a major landslide during Mitch has left a large area of unstable ground immediately above the

city. Figure 3 is an airborne image of a major

sediment deposit area on the Rio Choluteca

River where a major bridge was destroyed and

many homes were washed away. Figure 4 shows

the integration of the image information into a GIS database for use in planning reconstruction

and also to potentially form a ftiture framework

for the information infrastructure in this geographic area.

Because of the vulnerability of the city of

Tegucigalpa to additional flooding and

landslides during the rainy season and the

limited time available to put protection measures

in place; the majority of the initial effort in developing action plans was placed on short- term solutions to the flooding and landslides.

The effort to develop more long-term solutions

required substantial additional data gathering

and engineering analysis to determine the most

cost effective solutions. These studies and data

collection efforts have been underway and a

series of proposed approaches have been

developed. One of the focus areas is the

reconstruction sewage treatment system. The

report on this project should provide a plan

showing the preliminary layout of the sewage

collection system, pumping station locations,

and location/alternate locations for the treatment

plant. This effort would address alternate

516
treatment plant processes and the reason for selecting a particular process. The report would

be coordinated with the city officials and three

design review meetings would be held at Tegucigalpa. It was estimated that the report

would take 12 months to prepare. For this project topographic maps and existing sewer

system plans have been be reviewed and used

extensively in developing the alternative treatment plant and system options. The service area and potential locations of the treatment

plants have been developed and some

construction has been initiated, but the progress over the last year has been slow and is well

behind the original plan for reconstruction.

4.4 Summary of Initial Response Lessons

From Hurricane Mitcli

Hurricane Mitch was a natural disaster of epic

proportion to Central America. The recovery

and reconstruction efforts have also been of a magnitude not encountered before in that region.

For the first time in a major disaster, extensive

application of geospatial technology was used to

assist in all phases of the recovery. A digital atlas and Geographic Information System was

created for the impacted area to form a framework to integrate new data information over time in a systematic manner. Satellite,

aircraft, and ground information sources were

used to collect data and information to support

the recovery activities. A variety of products were successfully generated tailored to user needs in the field and higher-level decisionmakers in the recovery effort.

5. INTER AMERICAN DEVELOPMENT

BANK RECONSTRUCTION SUPPORT

The Inter-American Development Bank (IDB)

approved $597 million in financing for the

countries of Central America that suffered severe damage from Hurricane Mitch during the

one-year period immediately following one of

the worst natural disasters ever suffered by the

region. Simultaneously, at the request of Central

American governments, the Bank identified more than $280 million in resources for projects

that had been approved prior to the impact of

storm that could be converted into financing for new programs for disaster recovery. Disbursements to the region reached $540

million in support of reconstruction in Honduras,

Nicaragua, Guatemala and El Salvador. A major

portion of Bank reconstruction financing was

focused on social sector investments.

The IDB resources are part of a commitment by

the international community to provide more

than $9 billion in humanitarian assistance, longterm financing, and debt relief to countries

stricken by Mitch. IDB President Enrique V.

Iglesias reiterated that the institution expects to contribute about $3.4 billion over five years to the task of recovery and transformation of

Central America, a region with deep and

persistent social problems and a high level of

vulnerability to natural disasters. "A year ago

Central America's peoples suffered a catastrophe

of biblical proportions," Iglesias said. "Now they

are making progress on the monumental task of

rekindling development. Although the challenge

ahead is not something that can be resolved

quickly and easily, progress is being achieved in the context of democracy and macroeconomic

stability with the support of the international

community" (Inter American Development

Bank Web Page, 2000).

Hurricane Mitch unleashed over Central

America in less than a week the equivalent of one year of rainfall on a region parched by eight

months of drought. Its destructive force, made

worse by diverse environmental problems,

triggered massive flooding and landslides.

Nearly 10,000 persons died because of the

disaster and thousands more disappeared.

Millions of Central Americans lost their homes.

517
their land, and their crops. Infrastructure that had

represented decades of investment was swept

away. Economic losses rose to more than $5

bilHon, the equivalent of 17 percent of the

region's gross domestic product. The IDB

responded immediately to the Central American

crisis. The Bank sent technical missions to the

four countries most affected by the disaster in order to assist IDB country offices in evaluating

the damage, identifying available resources, and

preparing the most urgent counter-measures.

President Iglesias personally visited the region

to learn the priorities of the Central American

countries.

In the first few months following the destruction

of Mitch, the Bank concentrated its assistance on emergency programs covering a wide program

of activities, ranging from help in the

preparation of national plans for reconstruction

and transformation to support for social

programs. Among other projects, the Bank

supported rehabilitation of water and sanitation

systems, rural roads, bridges, schools, rural health clinics, and other high-priority social

services.

At the request of governments of the isthmus,

the IDB organized a meeting of donor countries

and multilateral organizations to form the

Consultative Group for the Reconstruction and

Transformation of Central America, an initiative to coordinate international cooperation. The

group met at the headquarters of the Bank in Washington, D.C. in early December 1998 to analyze the response to the emergency and the

preliminary studies, prepared by the United

Nations, on the damage caused by the hurricane.

The Consultative Group, chaired by the IDB,

met again at the end of May in Stockholm, with

support from the Swedish government. There the

Central American countries presented their national plans for reconstruction and

transformation. During that meeting the donor

community also had an opportunity hear the

opinions of representatives of diverse civil society organizations from Central America on

the impact of Mitch. At the end of the

conference, donors pledged more than $9 billion

in assistance, an unprecedented amount for this kind of aid effort.

The Central American recovery plans, whose

main documents can be accessed on the IDB's

website (www.iadb.org), are designed to reactivate the economies, generate productive

employment, reduce poverty, and promote the

sustainable use of natural resources. They also

seek to strengthen the democratic system of

government at all levels, encourage the

participation in government by the citizenry, and

improve the mechanisms of prevention and

mitigation of natural disasters.

In response to Mitch the IDB also organized

regional workshops attended by officials from

Central America and by experts from donor

countries on social and environmental

vulnerability, the reduction of the risk of natural

disasters, and emergency management measures. The Bank also sent several missions to the

region to discuss initiatives with the

governments to improve the transparency and

efficiency in the use of public funds, one of the

major concerns of the donor community.

The IDB put special emphasis on programs to alleviate poverty, rehabilitate basic

infrastructure, and protect natural resources. These projects are designed to improve the

efficiency of services and increase their coverage to include the most vulnerable social

groups as well as to encourage decentralization

and the participation by the private sector and

civil society organizations. One outcome of the

financial assistance effort by the IDB has been a

net positive flow of resources to the four

countries hardest hit by the hurricane. In keeping

with the Stockholm agreements, the Bank is due

to hold in February a round of consultations at

518
the national level with Honduras and Nicaragua

to analyze the progress in reconstruction and

transformation.

In November of 2000, the regional consultative

group will meet in Madrid at the invitation of the

Spanish government to discuss Central

American proposals. This forum will help

highlight the region's strategic advantages in terms of its geographic location, access to major

markets, and recent economic reforms designed

to stimulate trade and investment. At the same

time, a series of workshops will be held on

regional development. Among the issues that will be discussed are the benefits of furthering

Central American integration through road and

electricity projects and initiatives to reduce

environmental and social vulnerabilities by means of improved management of shared

watersheds and the establishment of early

warning systems for natural disasters.

6. PRIVATE SECTOR SUPPORT

ACTIVITIES

An important component over the overall

reconstruction effort in Central America has

been voluntary contributions and support from

private, commercial and charitable

organizations. Such support is usually

coordinated through international organizational

structures and the host country receiving the

support. Several illustrative examples are provided in this section.

6.1 Catholic Relief Services

On September 18, 1999 a bridge donated by J.F. White Construction in Massachusetts was

inaugurated yesterday in Moseli, Nicaragua, a small village 250 kilometers north of Managua

(Catholic Rehef Services Web Page, 2000).

Catholic Relief Services, which has an office in Nicaragua, worked with the construction

company to relocate the bridge from

Massachusetts to Central America. The bridge

replaced one that was destroyed one year ago by

Hurricane Mitch. The bridge spans the river

between the states of Jalapa and San Fernando

(Catholic Relief, 1999.

Following last year's hurricane, the

Commonwealth of Massachusetts and local construction companies, working with the

Archdiocese of Boston, joined forces and

donated 37 Bailey bridges and related equipment

to Catholic Relief Services for assembly in Honduras and Nicaragua. "Catholic Relief

Services is extremely appreciative of the efforts of the Commonwealth of Massachusetts, the

Archdiocese of Boston, and the numerous firms

and individuals that have helped to make this assistance happen," said Louise Wilmot, Deputy

Executive Director of Catholic Relief Services,

who attended the ceremony. "This bridge will not only reconnect the 60,000 people who were

left isolated since the hurricane destroyed the

bridge, it will also link the people of

Massachusetts and Moseli."

People in the rural region primarily support

themselves through the production of rice, com,

coffee, lumber, and cattle. Without the bridge,

they were unable to deliver their products to the

markets, resulting in great losses of income.

"This is an example of a public/private

partnership responding in an innovative and

important manner," commented Mark Snyder,

Catholic Relief Services' Country Representative

in Nicaragua. In Nicaragua, more than 100

bridges on primary and secondary roads were

destroyed by the storm leaving many areas completely isolated. "The technical support and

the bridges will make a tangible difference in people's lives here," Snyder said.

Catholic Relief Services is the official overseas

relief and development agency of the U.S.

Catholic community. Founded in 1943, today

519
the agency provides assistance to needy people

in more than 80 countries.

6.2 Environmental Systems Research

Institute (ESRI)

As part of the early support to Hurricane Mitch

ESRI provided multi-hazard maps and

information to emergency response

organizations, military, national and local governments. The data will serve as an

educational tool across the Central America by

sharing geographic knowledge about local hazards and prompting mitigation of those risks (ESRI Web Site, 2000). These same geospatial

based information resources can be used to assist

in reconstruction and building Disaster-Resistant

Communities. The focus is on promoting local preventive measures that will help reduce the

loss of lives and property during disaster events.

It is a commonsense approach based on three

principles—preventive actions must be decided at the local level, private sector participation is vital, and long-term preventive efforts and

investments are essential.

Some of the technology applications and

methods provided to Hurricane Mitch recovery

and reconstruction parallel current efforts in the

United States under Project Impact being lead by

the Federal Emergency Management Agency

(FEMA). FEMA is taking a proactive approach

to change the way the public deals with

disasters. Project Impact: Building DisasterResistant Communities is a nationwide initiative that FEMA launched in October 1997. It helps

communities protect themselves from the

devastating effects of natural disasters with

preventive actions that dramatically reduce

disruption and loss. The initiative focuses on

promoting local programs that will prevent

extensive loss of lives and property during

disaster events. It is a commonsense approach

based on three principles—^preventive actions

must be decided at the local level, private sector

participation is vital, and long-term preventive

efforts and investments are essential.

Local agencies have detailed, hands-on

knowledge of local issues, which makes them

ideally suited to facilitate disaster planning,

mitigation, and preparedness. And currently 120

communities throughout the United States are

participating in Project Impact. FEMA is also

recruiting other federal agencies and businesses

into the program to help get the latest technology and mitigation practices

implemented at the local level. Additionally,

Project Impact is serving to empower

communities as they pursue smart growth

strategies while strengthening the federal government's partnership with local government.

A common thread that weaves its way through

every phase of Project Impact is GIS. Most of

the data requirements for disaster management

are of a spatial or geographic nature, and that makes GIS the logical tool. Identifying potential

hazards, planning mitigation programs, and

assisting in disaster preparedness activities are some of the critically important roles GIS has

taken on for the Project Impact communities.

With GIS, local agencies can compare maps of

hazards such as earthquake faults, fire risk areas, or flood areas to other map data. Locating

streets, pipelines, hospitals, residential areas, power lines, and storage facilities in relation to known hazards on maps makes the process

seamless.

7. CENTRAL AMERICA

RECONSTRUCTION ACTIVITY

SUMMARY

Since Hurricane Mitch struck Central America,

project participants and local governments have

worked to help people recover from the storm's tremendous damage. The four Central American

520
countries hardest hit by Hurricane Mitch in 1998

have made immense progress. More than

21 million storm victims have been assisted

through programs in Nicaragua, Honduras, El

Salvador and Guatemala.

The first two phases of post-storm operations in the Central America emergency response and

rehabilitation are largely finished. In the next

phase, project participants will adapt long-term

development work to incorporate ongoing

recovery work that has at its core the promotion

of community organization, building capacity of

local governments, and the formation of a trust- based link between communities and their governments. This critical link is needed to achieve a sustainable recovery. Transportation

will be one of the focuse areas. There are many

remaining bridges and roads that need repair and

replacement throughout Central America (Figure

5). Simultaneously, capitalize will continue on

its efforts to strategize regionally across the four

Central American countries. From the

geophysical to the political, the social to the economic, Honduras, Nicaragua, El Salvador

and Guatemala share characteristics that should

be addressed across national borders (US AID

Web Site, 2000).

f There remains a commitment to the

i transformation of Central America over the long

I term, and programs will continue to work in

I collaboration with communities, local

i governments and local nongovernmental

organizations — as well as to advocate at the

j national and international levels ~ ways to

|i improve the lives and livelihoods of the

j, marginalized people of the region. A summary

' of activities completed and underway in each of

i the four countries are described in the following

' sections.

7.1 Nicaragua

Mitch-related activities are being transformed

into larger development programs, mirroring the approach used in neighboring countries. The

current work includes, categorized as sustainable

agriculture and environmental protection;

housing repair and reconstruction; temporary

employment; and water and sanitation, are summarized below:

Sustainable Agriculture and Environmental

Protection: In Matagalpa Department,

Agricultural Rehabilitation Project now largely

focuses on development rather than

rehabilitation activities, although farmers badly

affected by Mitch will continue to receive extra assistance until the end of the year. This project

helped over 3,000 households rehabilitate and

protect their agricultural and forested lands and

reestablish crop production. The Agriculture and

Environment Project, operational throughout

1999 in Leon and Matagalpa Departments,

helped 1,785 families diversify their food

production and reforest about 250 acres of land.

It is intended to extend this project for at least an

additional year, and link it to agricultural

rehabilitation and improvement in Posoltega.

There, work will continue through November

2001 with residents of Finca Santa Maria (see

housing repair and reconstruction, below) and

1,150 other families in the area to reestablish

crop and small livestock production, identify and

test alternative crops, and develop irrigation

schemes where feasible. The Seeds and Tools

Project (Esteli and Matagalpa) included Erosion

Control Project (Matagalpa) and La Esperanza

Project (Ciudad Dario).

Housing repair and reconstruction: The

Vivienda Esteli Project helped 142 homeless

families reconstruct their houses. By October, all 142 houses were completed, each with a water

tap and a sanitary latrine. Participants in the

Esteli Rural Infrastructure Project will soon

finish building a road to the small community,

after which residents, and the municipal

521
government of Esteli, will hold an official opening ceremony.

In October 1999, housing reconstruction projects were launched in La Trinidad and Posoltega. In both places, problems obtaining legal title to suitable land delayed project start-up. In La

Trinidad, participants will build 49 houses in a

jointly planned community. In Posoltega, 350

families will build homes over the course of the next two years. Figure 6 shows some of the

reconstruction activities underway. In most

cases reconstruction is being done above

historical high water lines. There are still some

ateas where expedient reconstruction is accuring

in the flood plain. The resulting Finca Santa

Maria community will include a school, health

facilities, green spaces and a water system.

Participants also have been engaged in ongoing

agricultural rehabilitation and a good local harvest from the first post-Mitch agricultural season was also very positive. Other

organizations will contribute counseling through

the agricultural training center and several small

economic activities project to the residents of the

future Finca Santa Maria.

Temporary employment: In Condega, Pueblo

Nuevo and San Juan de Limay municipalities of

Esteli Department, the Rural Infrastructure

Project created short-term employment

opportunities through cash for work projects. In October funding was secured for an additional

two years of project activities. During the

project's first phase, one or more members of more than 2,100 households in 60 communities

rebuilt 101 kilometers of road, cleared riverbeds

of flood debris, and built retention walls and

sewers.

Water and sanitation: The Water and

Sanitation Project in Matagalpa and Jinotega

that was recently transformed into a one-year

development project rehabilitated 65 water

systems and 204 wells serving a total of 35,880

people. In Chinandega, the El Viejo Water

Project will continue into 2000, providing

participants with new water systems and latrines and the training to maintain and repair them. The

Betesda Water Project in Posoltega reestablished

clean water for 800 people and built 92 latrines. As mentioned above, a water system forms part

of the ongoing work in Finca Santa Mario, and

also in the Posoltega area.

7.2 Honduras

There were seven Mitch-related projects in Honduras. Four of the projects are now

finalized, and two will conclude within a few

months. The seventh has evolved into a three- year development project, in line with a longerterm recovery strategy in the Central America

region. Among these projects are river channel

clearing and construction and stream bank

stabilization to reduce the threat of future

flooding. Figures 7 and 8 illustrate some of the

ongoing projects in this area.

Food for Work: The food for work program

began in December 1998 and ended in October

1999. The project encompassed agricultural

recovery, environmental rehabilitation, and

infrastructure repair and construction in 51

municipalities in 10 departments. In all, 36,000

families received basic food commodities ~

maize and beans — in exchange for their work.

An additional 2,000 pregnant or lactating

women and children under 5 received

supplemental food in the chronically food-poor

departments of Lempira and Intibuca. With tools and other relevant supplies participants

rehabilitated farmland, restored water and

sanitation systems, rebuilt bridges and retaining

structures along riverbanks and repaired houses

damaged by the storm. With the addition of

seeds and tools from many Food for Work

participants were able to again produce food for

their families. The objectives were to ensure the

short-term survival needs of families while

522
helping them restore their productive capacity

and support their economic revitalization. In all cases, Food for Work projects was three-party

agreements between project sponsors,

participants and municipal governments.

Construction and digging tools, seeds and

agricultural tools, and other items required by

communities in their rehabilitation and

construction work were provided This project

component was finalized in September 1999.

Resettlement of the Homeless: In early

October, heavy rains caused renewed flooding on the Choluteca River. Families were forced

from temporary housing into their newly- or

partially-built homes in Renacer Marcovia, a community plarmed and constructed for approximately 3,000 residents of the original

Marcovia village, which was destroyed by

floods a year ago. Today, 254 houses are complete and 226 houses are nearing

completion. The project had installed temporary

latrines and water tanks, which served until January 2000 when permanent sanitation

systems were built. Renacer Marcovia's

population has played an important role in planning the new village's public and residential

areas, and has provided all the labor for the

cement-block homes.

Reactivating Household Economies

(REMAM): As of late October, 177 community

credit committees in 150 communities had

disbursed loans averaging $195.60 to nearly

2,300 families. More than 70 percent of

borrowers had launched small-scale commercial

activities, and the remainder opted to begin

production or food processing businesses. The

REMAM project, an emergency economic

rehabilitation program, will continue through

January 2000, disbursing loans to a total of

3,000 families in six departments, including the

families participating in the Renacer Marcovia

project.

In the first phase of the Choluteca and Valle

Water and Sanitation project 18 communities

repaired or rebuild storm-damaged water

systems in eight municipalities of Choluteca and

Valle Departments. By the end of October, eight

of the water systems had been designed and

three systems were under construction. CARE

and municipal authorities surveyed and

registered as protected areas the fragile

watersheds that feed these systems and that provide water for 500,000 people. Watershed

management training was provided for municipal authorities, and a signed a agreement was made with the Honduran National Forestry

Authority to provide additional training and

support to the project. This water, sanitation and

environmental protection project will continue

until April 2002.

Debris and Sediment Management: The

cleanup, removal and disposal of debris and

sediment is still underway in the branches and

main stems of the major drainage basins. Much

progress has been made, but many areas still need attention in order to reduce the risk of

future flooding as shown in figure 9.

Hygiene Education for Urban Populations:

This work, finalized in September,

complemented the Tegucigalpa Child Survival

Project by training 800 community members and

health workers to deal with water and sanitation

issues in disaster situations, to administer water

committees, and to manage post-traumatic

shock. Nearly 1,500 primary school students

learned about health and hygiene through this project component, and residents of 24 urban

slums completed infrastructure projects,

repairing or building latrines, drainage canals

and retaining walls in their neighborhoods.

There remain many health and sanitation issues

throughout the country as shown in figure 10. In the case of the Villaneuva Water and Sanitation

Education Project it helped 1,275 families in

523
Villanueva municipality overcome water and

sanitation problems resulting from the hurricane.

Technical assistance was provided and materials

to build a 80,000 gallon water storage system,

the municipal government and other sponsors

installed the principal branches of the water

distribution system; and project participants were responsible for installing connections to bring water to their homes or communal taps.

Training representatives of the participating

communities in system maintenance was also

part of the overall process of community

organization and participation. Projects will

continue in Honduras to address water and

sanitation, environmental protection, sustainable

agriculture, education and child health.

7.3 El Salvador

Mitch-related emergency and rehabilitation

work in El Salvador has transitioned into two

longer-term development programs that will

assist people in three Eastern and western

regions of the country (360,000 people and

163,000 people, respectively) to manage clean

water systems, maintain basic sanitation

facilities, protect the environment and

rehabilitate agricultural systems over the coming

years. A third project in Usulutan helped 12

families rebuild houses damaged by floods last year. The development programs in El Salvador

will focus heavily on water supply,

environmental protection and sustainable

agriculture.

7.4 Guatemala

The last of the Mitch rehabilitation activities in Guatemala helped residents rebuild 19

suspension footbridges damaged or destroyed by

flooding. In all, these bridges provide access to 93 communities and about 4,000 families. Since

the hurricane struck, Mitch-related work in Guatemala has included emergency response,

water and sanitation system rehabilitation.

agricultural recovery, reforestation and

watershed protection in the Polochic river basin,

Alta and Baja Verapaz Departments. Ongoing

development work in Guatemala includes village

banks, health, agriculture and environmental

protection and education.

8. CONCLUSION

The opportunity to leverage technology in the geospatial area to assist in disaster management

has never been greater. There are many tools and processes currently in use that have not been

applied to assist in disaster management.

Examples include the application of

hyperspectral sensor data, use of digital

multispectral video, and geospatial 3-D flythrough analysis. All of these capabilities could

support and enhance the information available to the disaster management decision-maker.

There are exciting changes expected to occur

during the next few years in the few years in the

geospatial community. New aircraft and spacebased sensor systems are planned that will

revolutionize the spectral and spatial resolution

of data available for commercial applications.

There will be challenges, particularly in the data

processing and management area, because of the

extremely large spectral databases these new

sensors will generate.

The spectral library project also provides an

opportunity to expand cooperatively our global

understanding of the vegetative and geologic

spectral makeup of the earth. The baseline data

from this effort would be invaluable in conducting change analysis assessments

following a natural disaster, such as an earthquake or hurricane. Using some of the

analysis tools described earlier, the analysts and

decision-makers could virtually walk into the

disaster situation. Other spectral technologies,

such as fluorescence, could be used to precisely

524
locate very small underground gas and oil leaks

remotely. Within the next decade, there may be

the ability to better harness geospatial analysis

capabilities to help reach the vision for the

cooperative exchange of timely, relevant

information useful during all phases of disaster

management - to save lives and reduce economic

loss.

9. REFERENCES

Atlantic Hurricane Summaries for 1998, AccuWeather Service, April 1999

Burton, Ian, Robert W. Kates, and Gilbert F. White, The Environment as Hazard, 2d Edition,

New York, NY: The Guilford Press, 1993

Catholic Relief Program, Web Site,

www.crp.org. May, 2000

Collier, Vincent, Hurricane Mitch Summary

Report, U.S. Army Topographic Engineering

Center, March 1999

Digital Atlas of Central America, Version 1.0, Center for Integration of Natural Disaster

information, U.S. Geological Survey, February

10, 1999

Disaster Information Task Force Report,

Harnessing Information and Technology for Disaster Management, Global Disaster

Information Network, 1997

Environmental Systems Research Institute, Web

Site, www.ersi.com. May, 2000

Inter-American Development Bank, Web Site,

www.iadb.org. April, 2000

Operations Summary, Interagency Support to Hurricane Mitch Reconstruction, U.S. Army

Corps of Engineers, January 25, 1999

Roper, William, Geospatial Analysis Support to National Disasters, U.S.-Japan Wind

And Seismic Panel, International Conference,

Washington, DC, May 1998

Tobin, Graham A., and Burrell E. Montz,

Natural Hazards: Explanation and Integration,

New York, NY: The Guilford Press, 1997

United States Aid for International

Development, Central America Support and

Development Activities, Hurricane Mitch

Recovery, Web Site, http://www.info.usaid.gov. May 2000

World Disaster Report, Disaster Relief Task

Force, United Nations, March, 1996

525
Problem

Disaster Event and

\ Phenomenology ^

Disoster : Manogemeiiit

Figure 1: The Primary Elements of Disaster Management

526
Figure 3: Airborne Sensor Image of Choluteca River Sediment Deposits near Choluteca City
Many still not repaired

Temporary structures are common for pedestrian _

traffic I

Surface transportation is

a challenge for vehicles

in many parts ofthe area g

Figure 5: Temporary foot bridges and Damaged Bridge Infrastructure

Figure 6: Policy Variations in Resettlement and Relocation Practices

528
gure 7: Preparation for Future Flood Protection Through Channel Clearing and Construction
Figure 9: Debris and Sediment Management Challenges Along the Choluteca River

• Many public

sewer systems

not fully

repaired

• Good drinking

water

challenges

• Exposure risks ''^^

Figure 10: Sanitation and Drinking Water Exposure Risks

530
storm Surge Disaster by Typhoon No. 18 in 1999

by

Hiroyasu Kawai ^'

also founded in the Japan Society of Civil Engineers. However, unfortunately the typhoon

No. 15 in 1959. named Isewan Typhoon in Japan, generated large storm surge and high waves in Ise Bay. The coastal dikes were col- lapsed, and consequently about 5,000 lives were lost.

J Hokkaido Is.

ABSTRACT

On September 24, 1999, the t\phoon No. 18 attacked the west of Japan, and induced storm

surge and high w aves at the high astronomical

tide. According to the tide records, the inundation traces, the witnesses, and the numerical

simulation, the meteorological tide anomaly

became more than 3 meters in the northern in- nermost region of Yatsushiro-kai Sea, and be- came more than 2.5 meters in the western re- gion of Suo-nada Sea. Sea water pushed by the

storm surge overflowed and overtopped sea

walls, and inundated the houses and the farming lands. Some breakwaters were slided. and

parapets were removed from the bases. That was the most enormous storm surge disaster since 1959 in Japan.

KEY WORDS: Astronomical Tide

Inundation

Meteorological Tide Anomaly

Typhoon

1. INTRODUCTION

Japan has several large bays facing the Pacific

Ocean, as shown in Figure. 1, and some typhoons take routes through the Japan Islands every year. Therefore, the J^anese in the

coastal zone have been suffering from storm

surge disaster.

After the experience of the tremendous storm

surge disaster by the typhoon No. 13 in 1953,

the Sea Coast Act of 1956 was issued by the

Japanese Government to prevent such disaster. The Committee on Coastal Engineering was

a

^ Okinawa Is. Pacific Ocean

Figure 1 Maximum meteorological tide anomaly in principal bays in Japan between 1945 and 1998

1) Senior Research Engineer, Hydraulic Engineering Division, Port and Harbour Research Institute, Ministry of Transport;

3-1-L Nagase, Yokosuka, 239-0826, Japan

531
After these disasters, the countermeasures by

structures such as breakwaters, coastal dikes,

and tide gates were developed in the pnncipal

bays m Japan. As the result, such enormous

disasters have not been occurred since 1959

until 1998, although the meteorological tide anomalies shown in Figure 1 were recorded.

However, on September 24 in 1999, many

coastal facilities and houses in the western re- gion of Japan were damaged by the typhoon

No. 18. That was the most enormous storm

surge disaster in Japan since the typhoon No.

15 in 1959.

In this paper, the conditions of the typhoon No.

1 8 in 1 999, and the charactenstics of the storm

surge in Kagoshima Bay, Yatsushiro-kai Sea,

and Suo-nada Sea are reported.

2. TYPHOON NO. 18 IN 1999

Figure 2 shows the route of the typhoon No.

18 in 1999, in companson with the other pnncipal typhoons affecting the Kyushu region.

The typhoon was generated to the south of the Okinawa Island on September 19, and ap- proached the Kyushu Island on September 24,

The maximum instantaneous wind velocity of

66.2 mJs was recorded at Ushibuka near the

southern entrance of Yatsushiro-kai Sea. The

typhoon passed by Yatsushiro-kai Sea, with an atmosphenc pressure at the typhoon center of

940 hPa, and at a speed of 40 km/h, in the

early morning. The typhoon landed on Omuta

in the Kyushu Island, crossed the northern part

of the Kyushu Island, went across Suo-nada

Sea, and landed again on Ube in the Honshu

Island.

The principal typhoons which caused the large

meteorological tide anomalies around the Kyushu Island are listed in Table 1. The symbol pc

and V represents the atmospheric pressure at the typhoon center and the progression speed, near Yatsushiro-kai Sea, respectively. The typhoon No. 18 in 1999 was one of the strongest

typhoons in these several ten years. The progression speed was slightiy small.

The astronomical tidal range between the H.W.L. (mean monthly-highest water level) and

the L.W.L. (mean monthly-lowest water level)

is approximately 4 meters m Yatsushiro-kai

Sea. In the case of the typhoon No. 19 in 1991,

the meteorological tide anomaly became large

at the low astronomical tide. On the other

hand, in the case of the typhoon No. 18 m

1999, the meteorological tide anomaly became

large at the high astronomical tide, therefore

the total tidal level became much higher than

the H.W.L.

Figure 2 Routes of principal typhoons

affecting the Kyushu region

Table 1 Characteristics of principal typhoons

typhoon No.

pc

(hPa)

V

(km/h)

®No. 16, 1942 950 45

(DNo. 16, 1945 930 50

(E)No. 29, 1950 965 25

(DNo. 15. 1951 945 75

(DNo. 22, 1955 950 40

(DNo. 19, 1991 935 50

(7) No. 18, 1999 940 40

532
3. NUMERICAL MODEL OF STORM

SURGE

(1) Winds Caused by Typhoon

The distnbution of the atmosphenc pressure in a Wphoon is given by the Mayer's formula as follows.

po = pc-Ap expi (1)

where, pc is the atmosphenc pressure at the center of a t\phoon. A/? is the pressure differential, po IS the atmosphenc pressure at the point with a radial distance r from the center, and ro is the influence radius.

Assuming the above distnbution, the gradient

wind speed Ugr is given by — exp

Par — I

J (2)

where, / is the Coriolis coefficient, p a is the density of atmosphere.

The wind speed is actually reduced by about

30% and the wind direction is changed inward

by about 30 degrees, by the friction along sea

surface. The actual wind speed, namely manne

wind speed. Ui can be assumed as follows. u^=augr (3)

where, Ci is the parameter assumed to be 0.70

in this p^er.

The movement of a typhoon also generates the

winds of which speed Ui is given by

U^= ——V (4)

where, V is the progression speed of a typhoon.

The directions of the wind component and the typhoon progression are identical.

(2) Propagation of meteorological tidal wave

The propagation of the meteorological tidal wave caused by a typhoon is represented by

the non-linear long wave theory. The equation

of continuity and the equations of motion can be written by

d T) dM dN

+ + =0

dM ddx

dt

dt

=+fN-gD

D , D dpo

(5)

(6a)

dx

1

( X sx p W

dN d \ MN

dx[ D

p w dx

{ d^M

dt

d ;A^-' dy [D _

d-M

df

-fM-gD

( " -sy p w

d T] D dpo

p w dy

' d'-N

dy (6b)

I by) d'-N

dx-"- df

where, t is the time, x and y axes are taken in the horizontal direction, g is the gravity acceleration, p »' is the density of sea water, /?o is the atmospheric pressure on the water surface. Ah

is the honzontal eddy viscosity coefficient. D

is the water depth, M and N are the flux flows. They are defined by

D=A+ 77 (7)

M=J" udz

-h

ri N=J vdz

(8a)

(8b)

where, z axis is taken in the vertical direction,

and z=-h at the sea bottom, z=t] at the water

surface, u and v is the component of the velocity in the direction of the x and y axis respectively. Additionally, t sc and r sy is the tangential stress along the water surface, and r bx and z by is the tangential stress along the sea bottom, in the direction of the x and y axis re- spectively, defined by

T p aCoW^^l W?+Wy-^ (9a)

(9b)

533
T b.= -~^MyfM^' (10a)

zby=-^^Nyp^^- (10b)

where, Co is the resistance coefficient along

the sea surface, IVx and Wy is the wind velocity

at 1 0 meters above the sea surface in the direction of the X and y axis respectively, n is the Manning's roughness parameter.

(3) Finite difference method and grid mesh

In this paper, the governmental equations are solved by a finite difference method with a

stacked gnd mesh for space successive steps

and the leap-frog method for time successive

steps.

The regions shown in Table 2 are combined

with each other. The region 1 includes a part

of the Pacific Ocean and the East China Sea

near the Japan Islands. The stonn surge is re- produced dunng 48 hours, namely from 12:00 on September 22, 1999 to 12:00 on September

24. The time interval between successive steps

is 4 seconds.

4. FIELD INVESTIGATION

The tide records are the most reliable data to understand the storm surge. However, only one

tide station is available in Kagoshima Bay. A

few tide stations are available in Yatsushiro-kai

Sea, however there is no in the northern irmermost region, in which the large meteorological

tide anomaly gave the enormous disaster. Therefore, it is necessary to ask the witnesses

in the affected areas the maximum water level and the occurrence time, and to measure the

heights of the inundation traces.

In this paper, the field data are classified into four categories shown in Table 3 by the reliability. The actual tidal level should be near the

categones A and B, lower than the category C,

and higher than the category D.

Additionally, the damages of the breakwaters,

the sea walls, and the coastal dikes were also

investigated.

Table 2 Grid Mesh

region number or

oTiH ^1 iU. nnintQ |_/V./liilo space

intpr\/ci 1 iilLCl val Common

1 74 X 75 16 7 km

2 81 X 87 .*T ISJil Kagoshima

Bay

66 X 57 11 .O8 kmKill 4 72 X 96 0 .6 km

5 72 X 126 0.2 km

6 . 57 X 63 1.8 km

Yatsushirokai Sea

7 132 X 135 0.6 km

8 96 X 135 0.2 km

9 150 X III 0.2 km

10 81 X 96 0.2 km

11 114 X 120 1.8 km

12 186 X 114 0.6 km

Suo-nada Sea 13 120 X 195 0.2 km

14 213 X 111 0.2 km

15 102 X 102 0.2 km

Table 3 Classification of field data

category content A - tide record with no error

B

- trace of the maximum water

level unaffected by wave runup

and wave overtopping

C

- trace of the maximum water

level affected by wave nmup

- trace of the overtopped water

D

- tide record without the peak

tidal level

- trace of the maximum water

level due to tide inside a

partially shield building

- trace of the maximum water

level at the backward of the sea wall or coastal dike

534
5. STORM SURGE IN KAGOSHIMA BAY

Kagoshima Bay is approximately 70 kilometer

long from south to north as shown in Figure 3, and is much deeper than the other bays m

Japan as shown m Figure 4. The majonty of

the typhoons going through or passing by the Kyushu Island may go up north in almost parallel to the axis of the bay, therefore strong

wind blows from the bay entrance to the northem innermost region m the case that a typhoon

goes up north at the west side of the bay.

Figure 5 shows the profiles of the meteorological tide anomalies at the principal points. The

meteorological tide anomaly became the maximum between 3:00 and 4:00. Dunng the penod. the astronomical tidal level was near the

M.S.L. (mean sea level), and the total tidal level was near the H.W.L. (mean monthlyhighest water level)

.

distance along bay axis (km)

0 10 20 30 40 50 60 70

20

a 40

60

80

Figure 4 Longitudinal sections of pnncipal

bays in Japan

TOE

oc

nj

• Obs. —-Cal. ;

®Makurazaki :

(0E 1 oc

TO

0) 0

-1

01 23456 789 10 11 12 time (hr)

Obs. Cal.

@kagoshima:

01 23456789 10 11 12 time (hr)

Figure 3 Bathymetry of Kagoshima Bay

Figure 5 Profile of meteorological tide anomaly in Kagoshima Bay

535
0 10 20 30 40 50 km

I I I \ \ I

Figure 6 Maximum meteorological tide ano- maly m Kagoshima Bay

j A cal.

lomaly 1 :^ 0

® (D @

point

Figure 7 Comparison of meteorological tide anomaly in Kagoshima Bay

Figure 6 shows the distribution of the maximum meteorological tide anomaly in Kagoshima Bay. The maximum meteorological

tide anomaly at the bay entrance and in the

northern innermost region was around 0.5 meter and 1 meter respectively.

As shown in Figure 7, the meteorological tide anomaly reproduced by the numerical model

agrees with that observed at the tide stations. In the figure, only the point (D Kagoshima is in Kagoshima Bay The point ® Makurazaki

and @ Shibushi is in the west side and the east

side of the bay respectively.

In the Kagoshima Port, breakwater caissons were slided by high waves.

6. STORM SURGE IN YATSUSHIRO-KAI

SEA

Yatsushiro-kai Sea is approximately 60 kilo- meter long fi-om south to north, as shown in Figure 8, that is similar to Kagoshima Bay.

However, Yatsushiro-kai Sea is very shallow

compared with the other bays in Japan. In the

northern innermost region, the water depth is less than 10 meters as shown in Figure 4, and

the coast line is v-shaped as shown in Figure

8.

Figure 9 shows the profiles of the meteorological tide anomalies at the principal points

shown m Figure 8. There are two tide stations

at the point ® Misumi. At the points (D

Yatsushiro and (D Misumi in the northern in- nermost region of Yatsushiro-kai Sea, the first large peak was followed by the smaller second

peak. The first peak appeared at around 6:00 near the high astronomical tide. Consequently

the total tidal level became much higher than

the H.W.L. ( mean monthly-highest water

level)

.

Figure 10 shows the distribution of the maximum meteorological tide anomaly reproduced

by the numerical model. The maximum mete

536
orological tide anomal\' is about 1 meter at the southern entrance region. 1.6 meters around the point (B Yatsushiro, increases sharply in the northern innermost region. The meteorological

tide anomaly is estimated to be at least 3 meters at the point ® Shiranui.

As shown in Figure IL the meteorological tide anomaly reproduced by the numencal model

agrees with that observed m the tide stations,

and that measured by the field investigation.

01 23456789 10 11 12 time (hr)

Figure 8 Bathymetry of Yatsushiro-kai Sea

c

0) 0

-1

• Obs. Cal. 1

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• •

: <

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Figure 9 Profile of meteorological tide anomaly in Yatsushiro-kai Sea

537
0 10 20 30 40 50 km

®(D(D@(D(D(7)®(D@

point

Figure 11 Comparison of meteorological tide anomaly m Yatsushiro-kai Sea

® Ushibuka, (2) Mmamata, (B) Ryugatake,

® Himedo, © Yatsushiro, (D Kagami,

(7) Shiranui, (8) Misumi. (9) Kumamoto,

® Kuchmotsu

Figure 10 Maximum meteorological tide anomaly in Yatsushiro-kai Sea

Figure 12 Matsuai Section of Shiranui Town

538
Figure 12 shows the layout in the Matsuai

Section of Shiranui Town, the point®. There

is a coastal dike with a height of 7.0 meters above the C.D.L. (chart design level) and three basins for fishing boats are located in the nor- mal direction to the face of the coastal dike. The crests of the seawalls of the basins are 5.5 meters above the C.D.L. Many houses are built around the basins, however the ground levels of some houses are much lower than the crests of the seawalls.

According to the residents, the tidal level might

become higher than the crest of the seawalls

before 6:00 a.m. Consequently, huge amount of

sea water flew over the seawalls, and inundated

the houses in ten minutes or so. The area be- came like a pond. It was difficult for the resi- dents living on the low groimd to get aware of

the quick nsing of the tidal level, because it was dark and the view of the sea was obstruct- ed by the seawalls and the coastal dike.

Photo 1 shows one of the inundated houses.

Rice hull on the roof is a trace of the inundation. The inundation height aroimd this house

might be 2.6 to 3.3 meters. While most resi- dents living on the first stories of two-storied

houses tried to go upstairs, some residents in one-storied house broke the ceiling through and

climb on to the roof Unfortunately 12 people were drowned in the first story.

According to an eyewitness living beside the

Nakanishi Basin, the maximum water level might be approximately 7.0 meters above the C.D.L. It means that the meteorological tide anomaly was 3.9 meters. On the other hand,

the trace of the overflow could not be found on

the coastal dike.

Figure 13 shows the profiles of the astronomical tidal level and the meteorological tide anomaly reproduced by the numerical model. It

IS reproduced that the total tidal level became

higher than the seawalls of the basins before

6:00 and did not exceed the coastal dike.

Photo 1 Inundated house

8

q 5 o

i coastal dike

X---; seawall -- 5- A. i: tide anomaly

r^-T--*^: -------

astronomicat tide

l3 h 1 1 1 i 01 23456789 10 11 12 time (hr)

L

Figure 13 Profile of meteorological tide in Matsuai Section

The houses at the backward of the seawall in the Koyagawachi Section, Ryugatake Town,

the point ® were also inundated by about 1.2 maters. As shown in Photo 2, the concrete block wall was collapsed, and the windows on

the first stories of the house were completely

destroyed. According to the witnesses, the in539
undation began at about 5.00 a.m. and the crest of the seawall at the forward of the houses

might be submerged. It was one hour before

the inundation m the Matsuai Section.

Figure 14 shows the profiles of the astronomical tide and the meteorological tide anomaly

reproduced by the numencal model. The total tidal level might be much lower than the sea- wall. The incident wave height is estunated by wave hindcasting to be approximately 2 meters. The houses at the first line from the seawall were heavily damaged. Therefore, the area might be inundated mainly by wave overtopping. The witaess that the seawall was submerged might mean that waves with long periods overtopped dunng the storm.

Photo 3 shows the breakwaters in the Odo Port near tlie inundated houses. Tlie section between

A and B was collapsed into sea by tlie high

weaves and the high tidal level.

-i 5 —^-r-i-i. i... I.

0

'

'

' '3- : 1

: : : I

0 1 23456789 10 11 12 time (hr)

Figure 14 Profile of meteorological tide

in Koyagawachi Section

Photo 2 Houses in Koyagawachi Section

Photo 3 Breakwater in Odo Port

7. STORM SURGE IN SUO-NADA SEA

Suo-nada Sea is approximately 100 kilometer

long from east to west, as shown in Figure 15. The sea is surrounded by islands, and the axis

of the sea is almost normal to the routes of

most tvphoons ^preaching the sea. The water

depth in the west region is less than 20 meters.

Additionally, there are several small v-shaped

bays in the northern coast, and the water depth

inside the these bays is less than 5 meters.

540
10 20 30 40 50 km

Figure 15 Bathymetry of Suo-nada Sea

A OB XC AD- cal.(l)

-X. o X . I.

= 2 CB ^

03 -a ^

•cal.(2)

.

W - "1\x

1 1 1 1 1 ®(2)(D@®®(7)(D(D®®®®(S)®®®

point

Figure 16 Comparison of meteorological tide anomaly in Suo-nada Sea

® Kanda, (D Shin-Moji, (D Aohama, ® Chofli, (D Ozuki, ® Sanyo, ® Onoda, (D West Ube, (D East Ube, © Ajisu, ® Yamaguchi, ® Aio, ® Hofu, ® Shin-Nanyo, © Tokuyama, ® Yanai, ® Iwakuni

It is assumed in the numerical model that the wind speed ratios Ci and Ci are 0.7 and that sea surface friction changes the marine wind

direction by 30 degrees. Such assumptions are adopted in general and are applicable to Kagoshima Bay and Yatsushiro-kai Sea. However, the meteorological tide anomaly computed by the numerical model, described by the

dotted line in Figure 16, is much smaller than

that observed in the tide stations, especially in the western region of Suo-nada Sea.

Suo-nada Sea is long from east to west, and

the sea is surrounded by the islands with high

541
mountains. Therefore, the wind speed and di- rection estimated with the general assumptions

do not necessanly agree with those measured

by the wind gauges. By adjusting the wind

speed and direction, the meteorological tide anomaly described by solid line in Figure 16

is obtained.

Figure 17 shows the profiles of the meteorological tide anomalies at the prmcipal points

shown m Figure 15. There are two tide sta- tions at the pomt © Tokuyama. The meteorological tide anomaly became the maximum in the order of the points ® to ©, namely from

west to east.

Figure 18 shows the distribution of the maximum meteorological tide anomaly reproduced

by the numencal model. The maximum meteorological tide anomaly is large in the west re- gion of Suo-nada Sea, especially larger than

2.5 meters around the point (6) Sanyo.

In the coast of Suo-nada Sea, many coastal and

port facilities were damaged. For example in the artificial island off the Shin-Moji Port, the point (2), the seawalls were leaned and the par- apets were put out of joint, by the high waves

and the high tidal level, as shown in Photo 4. In the Yamaguchi Port, between the points ©

and ®, the coastal dike was coll^sed and

consequently the farmland was inundated with

sea water.

Photo 4 Seawall in the artificial island off the

Shin-Moji Port

3 i

- 2

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o

c

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A Obs.(2)

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®Tokuyama :

1 1 1

0 1 23456 789 10 11 12

time (hr)

Figure 17 Profile of meteorological tide anomaly in Suo-nada Sea

542
Il

543
Table 4 comparison of conditions in inundated ares

Yatsushiro-kai Sea Suo-nada Sea

Koyagawachi Matsuai Habu

Inundation start time 5.00 a.m. 5.50 a.m. 7.30 a.m. Astronomical tidal level (CD.L. . m) +2.2 +3.1 +3.5

Meteorological tide anomaly (m) 0.9 <3.9 2.5 to 3.0 Total tidal level (C.D.L . m) 3.1 <7.0 6.0 to 6.5 Crest height of breakwater (C.DL. , m) none 5.0 to 7.0 5.5 to 7.0 Crest height of seawall or parapet (C.D.L. . m) 6.0 5.5 5.9 to 6.5 Ground level at the backward (C.D.L.. m) >5.0 >3.2 >4.8

Maximum inundation height (m) 1.2 1.5 Death toll 0 12 0

Figure 19 shows the layout of breakwaters and

parapets in the Habu Fishing Port and the resi- dential area at the backward of them. According to the eyewitnesses, the tidal level rose rapidly around 7.30 a.m. The inundation height

at the backward of the parapets reached about

1.5 meters. It means that the maximum tidal level was between 6 to 6.5 meters above the C.D.L. (chart design level) and the maximum

meteorological tide anomaly was 2.5 to 3 meters. A few fishing boats went over the parapets and attacked the houses. When the tidal level was the maximum, the wind became

weak but the wave height was still large.

Table 4 shows the conditions of the inundated

areas in Yatsushiro-kai Sea and Suo-nada Sea. In the Habu Section, the inimdation began after the sunnse, and the ground level is not so low,

therefore no one was drowned there.

8. CONCLUSION

The mail conclusions of this paper are as fol- lows:

(1) By the typhoon No. 18 in 1999 gave enormous storm surge disaster especially

m Yatsushiro-kai Sea and Suo-nada Sea.

(2) In Kagoshima Bay, the meteorological

tide anomaly became large in the northern

innermost region, and the maximum value was about 1 meter.

(3) In Yatsushiro-kai Sea, the meteorological

tide anomaly became large in the northern

innermost region, and the maximum value was more than 3 meters. The Matsuai

Section was inundated by high tidal level due to storm surge

(4) In Suo-nada Sea. the meteorological tide anomaly became large in the western re- gion, and the maximum value was more

than 2.5 meters. Many parapets and coastal dikes were damaged, and the Habu

Section was inundated with sea water, i

REFERENCES

1. Kawai.H. and T.Hiraishi: Influence of Storm

Surge on Failure Probability of Breakwater,

Okushin Tsunami/ UJNR Workshop. Booklet of Abstracts, pp. 100-101, 1998.

2, Kawai,H.; Storm Surge in Ise and Mikawa

Bay Caused by Typhoon, Proceedings of the

31st Joint Meeting of United States-Japan

Panel on Wind and Seismic Effects, 13p.,

1999.

544
APPENDIX

545

Task Committee Reports

547

REPORT ON TASK COMMITTEE A

STRONG MOTION DATA AND APPLICATIONS

Date: 19 May 2000

Place: National Institute of Standards and Technology, Gaithersburg, USA

Attendees: Japan side Takashi Nagao (Chairman)

Izuru Okawa

PHRI

BRI

U. S. side Mehmet K. (^elebi (Acting Chairman)

Richard Olsen

Michael Blackford

USGS

ERDC, WES

NOAA/NWS

1. Objective and Scope of Work

To coordinate and promote sharing of strong motion earthquake data among researchers

and practicing engineers, and to develop techniques and exchange information for

evaluating the destructive effects of earthquake motion.

The scope of work includes:

1 . instrumentation,

2. recording, processing, and analyzing strong motion data,

3. engineering characterization of ground motion,

4. design applications, and

5. seismic zonation.

The activities of the Task Committee include:

1 . regular exchange of data and publications,

2. creating procedures for disseminating significant strong motion digital data with

regard for the rights and expectations of (a) owner, (b) the users of data and (c)

the earthquake engineering community,

3. planning and conducting T/C workshops and meetings, and

4. coordinating relevant research activities.

2. Accomplishments

549
During the 32"*" UJNR Panel on Wind and Se ismic Effects meeting, presentations made

related to the M=7.4 August 17, 1999 Izmit (Turkey) and M-7.2 November 12, 1999

Duzce (Turkey) and M=7.3 September 18, 1999 Chi-Chi (Taiwan) earthquakes revealed

the wealth of strong-motion data recorded during these events. The additional data sets will impact earthquake engineering and engineering seismology studies and practices for

years to come. These earthquake records enriched the data base for near- fault (<10 km)

records by tenfold and they exhibit very large velocities (>300cm/s) and permanent

displacements (>8M). These records are in addition to those from the 1994 Northridge

and 1995 Great Hanshin earthquakes which had a tremendous impact on various aspects

of the strong motion studies, including characterization of near field motion and site response. In addition, activities conducted by Task Committee A members during 1999

and 2000 are:

1. Close monitoring of the on-going changes and new proposals for seismic

guidelines and codes, including those of UBC97, NEHRP97 in USA and

ISO/WG67 in Japan and the new International Building Code 2000.

2. Contributions of the chairmen and members of Task Committee A to 12'*^ World

Conference on Earthquake Engineering (12WCEE; Jan. 2000; Auckland, New

Zealand).

3. Official incorporation of COSMOS (Consortium of Organizations for StrongMotion Observation Systems). This new organization is now planning to work on

specific workshops and standardization of strong-motion data bases. Efforts are underway to determine procedures for expanded international participation.

4. Completion of K-net through the efforts of Japan-side Task Committee A

members. Its data availability with ease and timely fashion should be a model to both U.S. and Japan strong motion programs for future development. Data is being disseminated with CD ROMs in both Japanese and English.

5. Beginning of implementation of a new ANSS (Advanced National Seismic

Systems) initiative in the United States under the leadership of the USGS. Before

end of FY2000 (September 30, 2000), with the funding provided by U.S.

Congress, there will be 80 new real-time stations in the United States (40 in

Northern California, 20 in Pacific Northwest and 20 in and around Salt Lake

City). A brochure describing this initiative was distributed to UJNR members. In

future years, this program will deploy additional real-time stations will further

help in improving shake maps for use by earthquake response managers.

6. Contributions of the task committee members to the organization of a workshop

on "Strong-motion Instrumentation of Civil Engineering Structures" funded by

NATO, NSF, USGS, and others (Istanbul, Turkey; June 2-5, 1999). Co-director

of the workshop, Dr. (^elebi, reported that the proceedings of this workshop is now being finalized.

7. A USGS Report on August 17, 1999 Izmit (Turkey) earthquake [USGS Circular

1 193] with contributions from panel members has been distributed to participants

of the 32"^^ Panel Meeting.

8. A USGS Report describing the aims of the ANSS (Advanced National Seismic

Systems) has been distributed to the participants of the 32"^ Panel Meeting.

550
3. Future Plans

The future plans include:

(1) Coordination and follow-up, when appropriate, of ongoing US and Japan

developments in strong motion recording programs and data dissemination in conjunction with activities of recently established committees and

organizations such as COSMOS. Realizing the importance of strong-motion

data to earthquake hazard mitigation, T/C A suggests the establishment of

links on UJNR Web page to the Web pages of major US and Japan agencies

where strong-motion data information are available without restrictions.

Coordinate these links also with Cosmos.

(2) Recognizing the increasing emphasis in the engineering and emergency

response communities on near-real-time data availability, T/C A will continue

to promote the use of, and exchange of information on,new developments in

real-time strong motion data acquisition, processing, and notification. This

activity will be in conjunction with the proposed TC on Seismic Information

Systems. In this context, the task committee will follow with interest the new

ANSS (Advanced National Seismic Systems) initiative that is being

implemented in the United States.

(3) Drs. Okawa and (^elebi proposed, and the committee accepted, the

organization of a

2"'' US-Japan Workshop on Soil-Structure Interaction, in Tsukuba, Japan, in March 200 1

.

(4) Development of recommendations for the Common Agenda, possibly in a workshop regarding US and Japan needs for strong-motion measurements in densely urbanized areas to significantly improve pubic earthquake safety.

(5) Encourage, where and when appropriate, follow-up of ongoing developments

in seismic hazard map and seismic guidelines to be generalized for worldwide

use.

(6) In coordination with Task Committee G, a joint workshop will be held on

Seismic Information Systems and Damage Detection. The time and venue of

the workshop will be in coordinated with the Third High Level Forum being

planned during Fall 2000.

551
Report of Task Committee B

TESTING AND EVALUATION PROCEDURES FOR BUILDING SYSTEMS

Date: May 16, 2000

Place: National Institute of Standards and Technology, Gaithersburg, MD

Attendees: U.S. Side-- H. S. Lew (Chairman) NIST

Japan Side- C. Minowa (Acting Chairman)

M. Teshigawara

H. Sugita

NIED

BRI

PWRI

1. Objective and Scope of Work

The objective of the Task Committee is to develop and recommend rational test procedures

and to collect performance data of the static and dynamic response of structures through both

laboratory testing of prototype structures and filed testing of structures.

The Task Committee:

(1) Plans and conducts workshops and joint meetings to identify research topics and

develop joint research programs

(2) Coordinates research projects carried out by various laboratories in the U.S. and

Japan. Facilitates publication of research results and implementation of findings

in codes and standards.

(3) Facilitates exchange of research personnel, the exchange of technical information

and the use of available testing facilities. (4) Develops uniform testing procedures, including loading history, for comparison

of results of tests carried out by various researchers, and for establishment of data

base.

(5) Develops guidelines for interpretation of test results to improve the design of

structures.

(6) Develops methodology for evaluation and interpretation of test results.

2. Accomplishments

(1) As part of the third phase of the U.S. Precast Seismic Structural Systems Program

(PRESS), the testing of the five-story full-scale frame was completed at the

University of California at San Diego in August 1999. The results of the testing

have been published in the PCI Journal, Nov.-Dec. 1999.

552
(2) The research program on Composite and Hybrid Structures (CHS) in Japan was completed in early 1998. The Japan Structural Consultant Association

has completed applications studies for implementing research results. They are published in the proceedings of the 12 World Conference on Earthquake

Engineering in January 2000.

(3) The research program on CHS in the U.S. was initiated in 1995 under the

sponsorship of the National Science Foundation. Testing of composite wall

systems have been completed.

(4) The sixth and final U.S.-Japan Technical Coordinating Committee meeting on CHS was held in March 2000. The dissemination and archiving of the research

results were discussed.

3. Future Plans

(1) Task Committees B and C to hold a joint meeting during the year 2000 to discuss future research plans in the coming decades.

(2) The joint workshop on test procedure, documentation, retrieval of test data, and

experimental facilities will be planned in the U.S., in conjunction with the 33''' Joint Panel Meeting.

(3) The Task Committee will consider the future U.S.-Japan Joint Research Program

using available testing facilities in both countries including the 3-D Full-Scale

Earthquake Testing Facility, which is being constructed at Miki, Japan by the

National Research Institute for Earth Science and Disaster Prevention.

(4) The task committee will study the feasibility of establishing a network for live

transmittal of experimental testing.

(5) The Task Committee will strongly encourage the CHS program coordinators to prepare reports summarizing the experimental and analytical research activities on CHS in the U.S. and Japan. This report will also recommend the application

and implementation of the results in the design and analysis of CHS.

553
REPORT OF TASK COMMITTEE C

HIGH PERFORMANCE STRUCTURAL SYSTEMS; DESIGN, EVALUATION,

RETROFIT AND AUTO ADAPTIVE MEDIA

Date: May 16,2000

Place: National Institute of Standards and Technology, Gaithersburg, MD, U.S.A.

Attendees: U.S.-side — H.S. Lew (Acting Chairman) NIST

Japan-side ~ Masaomi Tesigawara (Acting Chairman) BRI

HidekiSugita PWRI

Chikahiro Minowa NRI

1. Objective and Scope of Work

The objective and scope of work and the future plans of the Task Committee (c) are as

follows:

(1) Exchange information and plan, and conduct workshops on new design

technologies, evaluation technologies, repair and retrofit techniques.

(2) Study advanced materials and methods for new construction and those for

repairing and retrofitting existing structures. (3) Develop reliable condition assessment systems for new, existing and damaged

structures. (4) Develop and harmonize performance based structural design developed in each

country.

(5) Coordinate research projects on design, evaluation and improvement of structures

in the U.S. and Japan to minimize duplication and maximize benefits.

2. Accomplishments

(1) The 1^ U.S. -Japan Technical Coordinating Committee Meeting on Auto-Adaptive

Media (Smart Structural Systems) was held on January 7-8, 2000. Research

programs of both countries were discussed and planned. Before the meeting, the

U.S.-Japan Workshop on Smart Structure Systems was held January 6, 2000.

(2) Conducted the 6"" (final) U.S.-Japan Technical Coordinating Committee Meeting

on Composite and Hybrid Structural Systems in order to summarize research

products and to discuss utilization (Information, dissemination, publication and

archiving) of research products, March 2000, USA.

(3) Contributed to the 2"*^ Grantees Meeting U.S. -Japan coordinating Research in Urban Earthquake Disaster Mitigation on January 22, 2000 at Berkeley, CA to exchange the progress of research.

554
Future Plans

(1) Develop a reliable methodology for screening and analyzing wind and seismic

resistant capacity of existing building including advanced instrumentation

technology and expert systems for condition assessment of existing structures. (2) Develop a metric system for assessing structural performance and compile a database of advance materials/systems that have potential for improving structural

performance of new construction, for use in rehabilitating existing buildings, and

for the purpose of quantifying damage.

(3) Investigate new methods for achieving desired levels of structural performance

using advanced materials and systems.

(4) Continue to encourage participation of consulting engineers, universities, national

and local government agencies involved in instrumentation, evaluation, condition

assessment, and retrofit and strengthening of existing buildings for seismic and

wind resistance.

(5) Hold the 2"^^ U.S. -Japan Technical Coordinating Committee Meeting on AutoAdaptive Media (Smart Structural Systems) before the end of 2000 in U.S.A.

555
Report of Task Committee D

EARTHQUAKE ENGINEERING FOR DAMS

Date: May 16, 2000

Place: National Institute of Standards and Technology, Gaithersburg, U.S.A.

Attendees:

U.S. Side-- Robert Hall (Chairman) USAE

Japan Side— Yoshikazu Yamaguchi (Chairman) PWRI

1

.

Objective and Scope of Work

To develop technical insights into better understanding of the response of dams to seismic effects, the T/C (D) will plan, promote, and develop research initiatives to assist in assuring seismic safety and economical protective countermeasures against earthquake loading for these structures. The scope of work includes:

(1) Methods of analysis for seismic design of dams including outlet works.

Comparison of design methods and criteria between U.S. and Japan.

Development of "Design Earthquake Ground Motions" for analysis and evaluation of dams. Assessment of investigation and dynamic analysis methods as tools (modeling, calculation codes).

(2) Dynamic characteristics of dam construction materials and site conditions. Strength and deformation characteristics during earthquakes (concrete, soil and rock).

(3) For the analysis of observed behavior of dams and outlet works during earthquakes:

• Construct a database that contains measured ground acceleration and dynamic response of dams and outlet works during earthquakes, and other related information which is necessary to evaluate their seismic

behavior.

• Investigate the mechanism of damages due to earthquake loading.

• Analytically compare observed behavior during earthquake to the pre-earthquake-resistant design.

2. Accomplishments

(1) In the field of seismic design for embankment dams, Mr. Tomoya Iwashita, PWRI, finished a one-year (August

1998 - August 1999) researdi at University of California, Berkeley.

(2) T/C (D) conducted the Second U.S. -Japan Workshop on Advanced Research on Earthquake Engineering for Dams in Tokyo, Japan on 7-8 May 1999. The Workshop was attended by 13 U.S. participants and over ICQ

Japanese participants, and it served as a productive vehicle for technical exchange among scientists, engineers

and specialists. The technical sessions focussed on a variety of topics, including dynamic analysis and seismic

stability evaluation of concrete and embankment dams, site characterization, vibration testing, and laboratory

testing for determination of material properties.

(3) Mr. Tsuneo Uesaka , Mr. Takashi Sasaki and Masahiko Murase, PWRI, visited U.S. Army Engineer Research

and Development Center (ERDC) 13-14 March 2000 to discuss the future cooperative research works and to exchange the latest results of research / investigation and technical information about dam earthquake

engineering.

3. Future Plans

(I) Exchange of the results of research / investigation and technical information about dam earthquake engineering

is encouraged.

556
(2) Exchange visits to the institutes concerned, of scientists and engineers between U.S. and Japan, is to be

extended for the effective communications.

(3) In view of the importance of constructing a database for seismic evaluation of dams and outlet works,

cooperative research on the standardization for location and installation of strong motion seismographs, and

collection and processing of measured seismic data will be started. Dr. Enrique Matheu (Department of Civil and Environmental Engineering, Louisiana State University) will complete a six-week joint research with PWRl

engineers during this year at PWRI. Dr. Matheu will work with PWRl engineers in the initial implementation of

the database. This discussion will be usefijl in developing the proper format of the data, which will be obtained

from strong motion instrumentation of dams currently under construction by Ministry of Construction. During

the project, he will also work with PWRI engineers in preparing a technical paper on the comparison of U.S. and Japan design standard for concrete dam. A proposal to support this research will be sent by June 2000 to the Japan Science and Technology Agency.

(4) Dr. Michael K. Sharp (U.S. Army Engineer Research and Development Center) will complete a four-week joint research with PWRl during the next year at PWRl to prepare a technical paper on the comparison of U.S. and

Japan design standards for embankment dam. The U.S. Army Corps of Engineers through the Earthquake

Engineering Program will fund this research effort.

(5) The proceedings of the Second Workshop will be published and distributed to the T/C members and Panel

secretariats.

(6) T/C (D) will hold the Third U.S. -Japan Workshop on Advanced Research on Earthquake Engineering for Dams

next year in USA. The location and time of the workshop will be determined through correspondence between

the chairs of T/C (D). Joint presentations of cooperative research mentioned in (3) will be made at the workshop.

\

557
REPORT ON TASK COMMITTEE E

DESIGN FOR WIND AND WIND HAZARD MITIGATION

Date: May 16, 2000

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S.-side -- Harold R. Bosch (Acting Chairman) FHWA

Japan-side

Hiroshi Sato (Co-Chairman)

Shoichi Saeki

PWRI

PWRC

1. Objective and Scope of Work

To exchange technical information and to jointly plan, promote and foster research and

dissemination, to improve understanding of wind and its effects on structures, establish more

rational wind resistant design methods for structures, and to contribute to wind hazard

mitigation.

The scope of work includes:

(1) Characterization of strong wind, especially boundary layer extreme winds.

(2) Wind effects (wind loading on and wind-induced response of structures).

(3) Experimental and analytical methods to predict wind and its effects.

(4) Damage and risk assessment.

(5) Wind hazard assessments and wind hazard mitigation.

2. Accomplishments

(1) Held the second Joint Workshop on Design for Wind and Wind Hazard Mitigation at the

PWRI on May 24-26, 1999. Twelve US-side Task Committee members and 15 Japan-side

members participated in the Workshop. After presenting papers covering wind

characteristics; wind flow, pressure and loading; wind-induced response; and wind hazards;

the participants separated into three groups: wind characteristics and wind hazards; wind

effects on buildings; and wind effects on bridges, and identified research topics of mutual

interest. Before and after the Workshop, the participants visited the following technical sites: wind tunnels of Meteorological Research Institute, BRI and PWRI; and the Honshu-Shikoku

bridges between Onomichi and Imabari.

558
(2) The proceedings of the second Workshop were published by PWRI, and distributed to all the

Task Committee members.

3. Future Plans

1 ) Plan the third Workshop in the autumn of 200 1 , possibly at the University of Washington in

the US.

2) Pursue the possibility of collaborative research on the following topics recommended at the

second Workshop.

a) In the field of wind characteristics and wind hazards,

(1) Wind effects due to complex terrain

(2) Modeling and validation with full-scale data

(3) Joint field studies

(4) Cooperative quick-response post-storm damage assessments

b) In the field of wind effects on buildings,

(1) Development of realistic aerodynamic loads for performance-based structural

design

c) In the field of wind effects on bridges,

(1) Improvement of methods for predicting wind-induced response of cablesuspended bridges with edge girders

(2) Prediction and mitigation of wind-induced vibration of stay cables

3) Exchange technical information on the following subjects:

( 1 ) Topographical effects on wind and the use of field measurements, numerical

models and geographic information system (GIS) to study these effects.

(2) Control of wind-induced response of structures.

(3) Computational fluid dynamics and wind tunnel tests.

(4) Prediction of wind-induced response of full-scale structures.

(5) Investigation of wind hazards, including post-storm damage assessment

methodologies.

(6) Wind resistant design codes, standard and recommendations.

559
REPORT ON TASK COMMITTEE F

DISASTER PREVENTION METHODS FOR LIFELINE SYSTEMS

Date: May 16, 2000

Place: National Institute of Standards and Technology

Attendees: U.S.-side- James Cooper (Acting Chairman)

Josephine Malilay

FHWA

CDC

Japan-side Keiichi Tamura (Acting Chairman) PWRI

1. Objective and Scope of Work

To improve the performance of lifeline systems during earthquakes and extreme winds, and to promote the development and implementation of technical and non-technical countermeasures,

including the capability in damage estimation techniques and inspection procedures, through:

3. Planning and conducting workshops.

4. Facilitating exchange of technical information and personnel.

5. Promoting development of design guidelines and standards.

2. Accomplishments

4) Members of this Task Committee (T/C) participated in the planning and attended the Second

High-Level U.S. -Japan Earthquake Policy Cooperation Forum held in Yokohama, Japan in

November 1999. The Forum was led by the U.S. Federal Emergency Management Agency

and the National Land Agency of Japan, under the U.S.-Japan Natural Disaster Reduction

Initiative of the U.S.-Japan Framework for New Economic Partnership (Common Agenda).

3. Future Plans

(1) Encourage collaborative research and development in areas such as: performance prediction

and post-earthquake damage assessment of lifeline systems; systems approach to lifeline

performance; vulnerability inspection and retrofit techniques; post-earthquake response;

and socio-economic impacts of damage to lifeline facilities.

(2) Encourage and strengthen current efforts in both countries for developing seismic design

guidelines and standards for lifeline systems. Existing UJNR channels should be fully

utilized to facilitate the exchange of relevant information concerning the development of

guidelines and standards. Possible collaboration of developing guidelines and standards for

560
lifeline systems should be pursued.

(3) Continue to explore cooperative opportunities to address post-disaster public health issues as a result of disruption to lifeline systems as a part of the T/C F. T/C F members will make

recommendations on the appropriateness of incorporating this topic within the T/C F scope

after the appointment of the U.S.-side Chair.

(4) Upon appointment of the U.S.-side Chair, plan and conduct the Eighth Joint Workshop on

Disaster Prevention for Lifeline Systems in Japan. Specific location, time, and theme of the

workshop will be determined through correspondence between the co-chairs of this task

committee.

561
Report of Task Committee G

SEISMIC INFORMATION SYSTEMS

Date: 17 May 2000

Location: National Institute of Standards and Technology

Gaithersburg, MD U.S.A.

Attendees U.S. side William Roper (co-chairman)

Stuart Nishenko (co-chairman)

Mehmet Celebi

(NIST)

(FEMA)

(USGS)

Japan side Hideki Sugita (co-chairman)

Izuru Okawa

Akihiro Sanada

Masaomi Teshigawara

(PWRI)

(BRI)

(PWRI)

(BRI)

1. Objective and Scope of Work

The objective of the Task Committee is to plan and coordinate collaborative activities

in the following areas:

1 . Technically assist the earthquake policy cooperation under the U.S.-Japan

Common Agenda for Cooperation in the Global Perspective (Common Agenda).

2. Improve understanding of earthquakes and social impacts through development of

seismic information systems.

3. Review the principles, objectives, structures, and methodologies of existing

seismic information systems and facilitate practical application in both countries.

These objectives will be achieved through a) plamiing and conducting workshops, b)

planning and conducting cooperative research, c) exchange of information and

personnel in administrative and technical fields, and d) introducing research

accomplishments to guidelines and standards.

2. Accomplishments

1. The T/C G members participated in the planning and attended the Second High

Level U.S.-Japan Earthquake Policy Cooperation Forum (High-Level Forum) that

was held under the auspices of the Common Agenda, 10-12 November 1999 in

Yokohama, Kanagawa, Japan. The High-Level Forum was led by the NLA and

FEMA, and political efforts for the earthquake disaster reduction including the

application of seismic information systems were discussed. Recognizing the

importance of continuous information exchange, both countries agreed to hold the

Third High-Level Forum in 2000.

2. The First U.S.-Japan Workshop on Seismic Information Systems was held at PWRI during 15-16 November 1999. 38 persons including 5 persons from the

U.S.-side and 18 persons from the Japan-side participated, and 27 papers were

presented during the workshop. Following the workshop, a technical site visit

562
was conducted at the NLA, the Tokyo Fire Department and the Tokyo Gas

Company, Ltd.

3. The Proceedings of the First U.S. -Japan Workshop on Seismic Information

Systems are pubhshed and will be distributed to the workshop participants and the

Secretariat.

Future Plans

1 . In consideration of the specific topics to be discussed in the Third High-Level

Forum, the Task Conmiittee will encourage collaborative activities in the

following areas: • Application of advanced technologies including GIS, GPS, and satellite imaging.

• Strategies to collect, organize, archive, and distribute data from individual

information systems.

• Cost-effective technologies for the application of seismic information systems.

• Information systems for tsunami hazard and risk reduction.

2. The Task Committee will pursue the possibility of collaborative research to compare loss estimation methodologies, information system architectures, and

risk assessment/crisis management strategies in both countries.

3. The Task Committee will pursue information exchange with other related UJNR

T/Cs and Panels on common technologies including earthquake ground motion

acquisition, remote sensing of earthquake damage, and tsunami hazard and risk

reduction.

4. The Task Committee will conduct the Second U.S.-Japan Workshop on Seismic

Information Systems at the USGS facilities in Menlo Park, California in 2000.

This workshop will be held in cooperation with Task Committee A and will

include damage detection in the agenda. The U.S.-side will propose a date for the

workshop that is in coordination with the Third U.S.-Japan High-Level Forum.

563
Report of Task Committee H

SOIL BEHAVIOR AND STABILITY DURING EARTHQUAKES

Date: May 17, 2000

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. -side— Richard Olsen (Acting Chairman) WES

Japan-side — Keiichi TAMURA (Acting Chairman) PWRI

Chikahiro MINOWA NIED

1. Objective and Scope of Work

Government agencies responsible for pubHc works must assure seismic safety and

provide economical protection against earthquake hazards.

The objective of the Task Committee (H) is to assist in meeting these needs by

enhancing the availability of technology for predicting the dynamic behavior of soils,

foundations and earth structures, and analyzing dynamic soil-structure interaction to assure their safe performance during earthquakes.

In accordance with the objective, the scope of work includes:

(1) Exchange information on technological developments, state-of-the-art and

practice related to soil behavior and stability during earthquakes,

(2) Exchange information and technical data related to field performance, research,

and methods of practice,

(3) Plan and conduct programs of cooperative research and/or workshops in

coordination with the proposed or ongoing programs, and

(4) Make other efforts needed including exchange of researchers between U.S. and

Japanese research institutes, and publication of research results and

recommended practice.

2. Activities

(1) The National Research Institute for Earth Science and Disaster Prevention

(NIED), the Building Research Institute (BRI), and Wayne State University

(WSU), have continued to work in their cooperative research program "Physical

and Numerical Simulation of Structural Damages Due to Liquefaction and

Development of Countermeasure Techniques", 1994-2004.

(2) Port and Harbor Research Institute (PHRI), Colorado School of Mines and the

New Mexico Resonance Institute (Albquerque, NM) have continued joint

564
research on application of the Nuclear Magnetic Resonance Imaging (NMRI)

method to the study of soil behavior and stability during earthquakes.

(3) Under the auspices of Task Committee "H", Mr. O. Matsuo and Dr. M. Okamura,

PWRI researchers in geotechnical earthquake engineering, had a successful visit

in August 1999 and December 1999, respectively, to WES and other universities.

The purposes of their visit were to collect recent information on evaluating

liquefaction potential, experimental techniques of geotechnical centrifuge

modeling and related topics, and to discuss possibility of developing cooperative

research works.

3. Future Plans

The U.S. and Japan sides will put a focus on exchanging research findings and

developing cooperative research works in the following two particular areas in which

both Japanese and U.S. researchers are currently conducting the studies.

(1) Seismic stability assessment and design procedures for embankment structures

Japanese researchers are investigating the rationalization of conventional pseudostatic stability analysis procedures and developing predictive procedures of

permanent deformation and flow failure.

U.S. researchers are investigating different types of failure mechanisms and the

effectiveness of current procedures for liquefaction analysis. Improvement of

centrifuge experimentation on embankment structures is a common interest.

(2) Improvement of liquefaction potential assessment procedures

Both the U.S. and Japan currently have procedures for evaluating liquefaction

that incorporate many factors. Both sides are currently involved in research to

evaluate these factors. In-situ measurement technique by the use of the cone

penetrometer is also of common interest for both sides.

The committee also plans an exchange of researchers in the coming year to exchange

information and encourage cooperative research activities.

565
Report of Task Committee I STORM SURGES AND TSUNAMI

Date: May 17, 2000

Place: National Institutes of Standards and Technology, Gaithersburg, Maryland, U.S.A.

Objective and Scope of Work

The objective of this Task Committee is to mitigate damage from storm surge and tsunami

through cooperative research and shared technology and information. The primary cause of

storm surge is considered to be tropical cyclones (hurricanes, typhoons). The primary cause of

tsunami is considered to be sudden sea floor deformation due to earthquakes, landslides, and

volcanic activity. Both hazards may cause disasters along coastal regions.

The scope of work of this Task Committee is as follows:

(1) Exchange results of research on storm surge and tsunami occurrence, generation,

propagation, and coastal effects. This includes observations on historical, current, and

theoretical tsunamis. Of particular interest is the effort by the U.S. and Japan to acquire

deep ocean tsunami measurements.

(2) Exchange results and status of storm surge and tsunami activities including analysis of

the problem, plarming, warning, and engineering approaches.

(3) Exchange information on the development of technologies such as computer programs

to predict travel times, landfall locations, wave characteristics, inundation and run-up

heights, improved instrumentation, and the use of satellite communication for detection

and warning.

(4) Facilitate dissemination through exchanges of literature, technical reports at joint

meetings, special workshops, joint projects, and direct interaction among participants.

The storm surge research community, which functions through many related societies

and international organizations, defines and stimulates work in the field. The Task

Committee, through its meetings and workshops, encourages exchanges of ideas and

joint study by U.S. and Japanese investigators of tsunami events throughout the world.

Accomplishments

Since the issuance of the last report of this Task Committee U.S. Side members have been active

in projects that have resulted in the publication of a number of reports of value to emergency

managers and the general public. Two reports of particular interest are, "1997-1999 Activities of

the Tsunami Mitigation Subcommittee - A Report to the Steering Committee, National Tsunami

Hazard Mitigation Program" and "Surviving a Tsunami - Lessons from Chile, Hawaii, and

Japan", U.S. Geological Circular 1 187. Through efforts of the Pacific Tsunami Warning Center

and the UNESCO/IOC Tsunami Warning System in the Pacific two new water level stations

have been installed and placed in operation in Kamchatka and in the northern Kuril Islands.

Attendees: U.S. -side —

Japan Side —

Michael Blackford, Chairman

Shigeki Unjoh, Acting Chairman

NOAA

PWRI

566
Data from these stations are relayed through the Japanese GMS satelHte to users in Japan, the

U.S.A., and Russia. Improved communications methodologies have vastly increased the amount

of both seismic and water level real time data available to the Warning Centers. These data help

to reduce the time necessary to both issue and cancel tsunami warnings and watches and to reduce the possibility of the issuance of unnecessary warnings.

Future Plans

The Task Committee exchanges information and encourages cooperative research and

development on the following subjects:

(1) Cost reduction of deep ocean tsunami detection systems

(2) Accessibility of tsunami detection data in the deep ocean

(3) Investigation of tsunami detection by artificial satellites

(4) Accessibility of gridded bathymetry data in the deep ocean and extension of the database

to shallow seas

(5) Development and improvement of numerical models of tsunamis and storm surges

(6) Development of predictive models on scour around structures by tsunamis

(7) Technical support to develop tsunami mitigation programs in Pacific nations

(8) Collecting information (tsunami source, run-up height and damage) for historical and

current tsunami in the Pacific

Other

The Task Committee shall hold its Sixth Tsunami Workshop through its joint sponsorship of a tsunami workshop hosted by the NOAA/Pacific Marine Environmental Laboratory to be held in

the U.S.A. in 2001.

567
Report on Task Committee J

WIND AND EARTHQUAKE ENGINEERING FOR TRANSPORTATION SYSTEMS

Date: May 17,2000

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. -side -- Dr. Phillip Yen (Acting Chairman) FHWA

Japan-side — Dr. Hiroshi Sato (Acting Chairman)

Dr. Kazuhiko Kawashima

1. Objective and Scope of Work

Surface transportation systems play a vital role in the movement of goods and people. Highway

bridges are especially influenced by the forces of wind and earthquakes because of their open

exposure to those forces.

The objectives of work includes:

(1) To plan, promote and foster research on the behavior of highway bridges when subjected

to wind and seismic forces, and

(2) To disseminate research results and provide specifications and guidelines based on the

Task Committee's findings.

The scope of work includes:

(1) To focus research on highway bridges without limitation on their size and function, and

(2) To investigate existing and new bridge design, the behavior of whole bridge systems

and/or single component of a bridge.

2. Accomplishments

The 15"^ U.S.-Japan Bridge Engineering Workshop was held during November 9- 10, 1999, in Tsukuba, Japan. The workshop was attended by 22 U.S. and 65 Japanese participants. The

earthquake reports of the Turkey Earthquake and Taiwan Earthquake were held as the special

session. The topics of the workshop included: 1) Innovative seismic retrofit technology including

the application of new materials and new structural systems, 2) Innovative bridge design concept,

e.g., bridges types, materials, life cycle cost analysis. The proceedings of the workshop have been

published and will be distributed.

3. Future Plans

(1) The 16"" U.S. -Japan Bridge Engineering Workshop will be held in October 2000 at Lake

Tahoe, Nevada, U.S.A.

PWRI

Tokyo Institute of

Technology

568
The topics of the workshop will include:

a) Corrosion protection technology of steel and steel materials in concrete.

b) Technology of performance based seismic design and innovative seismic retrofitting

for existing bridges.

(2) Encourage coordinated researches on the following topics:

a) Bridge design methodologies which embrace performance based design concepts,

limit state design concepts and life cycle costs analyses.

b) Seismic isolation design methodologies considering the dynamic characteristics of

isolation devices and their modeling.

(3) Continue to investigate and exchange technical information on the improved seismic

retrofitting and strengthening procedures for highway bridges based on experimental,

analytical, and field studies, including the development of innovative earthquake protective

systems, the evaluation of the seismic performance of total bridge systems, and the

development of damage evaluation methods using sensor systems, etc. This exchange

should also include information related to the maintenance of existing bridges.

(4) Continue coordinated research project to compare the seismic design criteria and seismic

retrofit technologies for bridges in Japan and the U.S.

(5) Encourage research study on seismic and aerodynamic response of long span bridges such

as suspension bridges and cable-stayed bridges with emphasis on behavior of composite

materials, cable inspection, vibration control, and corrosion protection.

(6) Encourage coordinated research study on system identification techniques, nondestructive evaluation of bridge structures, use and performance materials including new

materials, and performance ofjointless bridges.

569
Report on Task Committee K

WIND AND EARTHQUAKE ENGINEERING FOR OFFSHORE AND COASTAL

FACILITIES

Date: May 16, 2000

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. -side — Mehmet (^elebi (Acting Chairman) MMS

Japan-side ~ Takashi Nagao (Acting Chairman) PHRI

Yoshikazu Yamaguchi

1 . Objective and Scope of Work

To develop technical insights necessary to mitigate damage to offshore and coastal facilities due

to extreme wind and seismic effects. The Task Committee will plan, promote and develop

research initiatives to meet this objective and will disseminate the results of its research for

incorporation into future specifications and design guidelines. Criteria for the design of offshore

and coastal facilities differ from their onshore counterparts. These differences arise due to the

unique design or mass distribution of the facilities, to the fluid/structure or wind/structure

interaction, to the placement of foundation elements in or on soft, fully saturated soils that can be

subject to large hydrodynamic pressures, and to the lack of specific environmental data or engineering experience that has been developed for most Onshore Sites.

The scope of work includes:

(1) Sponsoring and conducting workshops and meetings to identify key areas of research

opportunities for cooperation and the exchange of knowledge.

(2) Predicting strong motions for offshore and coastal sites including assessing the effects of

basin geometry, and linear and nonlinear local geological effects, using actual sea-floor

response measurements.

(3) Determining the dynamic response and the interaction of structure/foundation/soil systems

to seabed motions and extreme wind forces.

(4) Assessing the dynamic response and behavior of various operational facilities mounted on

offshore and coastal structures.

(5) Developing assessment methodologies for earthquakes and other characteristics of

potential seismic sources (e.g. faults) for offshore and coastal sites in regards to how these

conditions relate to structural design criteria. (6) Promoting the implementation of new research results into current design and construction

processes.

(7) Developing research efforts to include laboratory and field programs to obtain data on the

response of offshore and coastal facilities to extreme wind and seismic forces.

(8) Creating performance standards, design specifications, and code recommendations for

570
(9) applications to new construction as well as remedial action for existing facilities.

2. Accomplishments

(1) Task Committee K held initial planning phase meetings to conduct a third U.S.-Japan

workshop. The theme of this workshop will focus on seismic, wind and hazard mitigation

techniques for offshore and coastal facilities. (2) Mr. Nozu, researcher of PHRI, completed research on earthquake resistant technology for

offshore and coastal facilities at California Institute of Technology in September 1999.

3. Future Plans

(1 ) Develop and finalize workshop agenda including invited speakers of the third U.S. -Japan

Workshop on Wind and Earthquake Engineering for Offshore and Coastal Facilities. (2) Coordinate, where possible, on-going research on wind and earthquake engineering for

offshore and coastal facilities of interest to the members of the Task Committee K. This

includes sharing of research reports and publications, where possible, listed as following.

a) Seismic performance based design concept for onshore and offshore facilities. b) Rehabilitation/life extension technologies for offshore and port facility.

(3) Task Committee K will work within the fi-amework of the U.S.-Japan Common Agenda

Program focusing on the issue of acceptable seismic risk guidelines for ports and harbors.

(4) Task Committee K will coordinate activities with the Pacific Earthquake Engineering

Research (PEER) Program on ports and harbors centered at the University of Southern

California.

571

Technical Site Visits

573

SUMMARY OF THE TECHNICAL SITE VISITS 20-23 MAY 2000

Following the 32"'^ Joint Meeting of the UJNR Panel on Wind and Seismic Effects, a joint

delegation traveled to Puerto Rico to visit four technical sites. Puerto Rico is a self- governing commonwealth of the United States. The delegation visited sites in San Juan

and Arecibo in the northern part of the island and traveled to the city of Ponce on the

southern coast to visit two sites. The details of these visits are described below.

1. Arecibo Observatory

The delegation visited the Arecibo Observatory on May 2 1 . The Arecibo Observatory is a part of the United States' National Astronomy and Ionospheric Center (NAIC). Cornell

University operates the observatory under a cooperative agreement with the National

Science Foundation, with additional funding for its operation provided by the National

Aeronautics and Space Administration. Mr. Jose Maldonado hosted the delegation and

led the tour of the facility.

The Arecibo radio telescope is the largest single dish radio telescope in the world. The

dish measures 305 m in diameter and 51 m deep. The spherical reflecting surface is composed of approximately 40,000 1 m X 2 m panels supported by a network

of steel cables. The observatory is located in a large limestone sinkhole in

the Karst region near the city of

Arecibo. The observatory is used to conduct experiments in the ionosphere,

radar studies of nearby planets and

asteroids, and to study radio emissions

of pulsars and quasars.

The antenna platform is suspended

137 m above the surface of the dish.

The 800,000 kg platform is suspended

by cables strung from three reinforced

concrete towers. Two of the towers

are 80.8 m tall, and the third is 1 1 1 m

tall. The differing heights were

necessary so that the tops of the towers were all at the same elevation. Each

tower is back-guyed to ground anchors

with a system of seven 8.26 cm steel cables. A separate system of cables

connects each corner of the platform to large concrete blocks under the dish.

The concrete blocks are attached to jacks, which permit adjustment of the

575
Figure 2: One of three reinforced concrete towers supporting antenna platform. The

control building is visible at the lower left and

the Visitor Center can be seen at the base of the tower.

precision reflector was installed during an up:

planetary radar. The gregorian dome was ins

1997.

height of each corner of the platform with

millimeter precision. The triangular

frame of the platform supports a circular

track from which the azimuth arm is suspended. The circular track allows for

the rotation of the azimuth arm in the

horizontal plane. The azimuth arm itself

is a 1 00 m long structure that includes a curved track. On one side of the track is mounted the gregorian dome and on the

other the carriage house to which a number of linear antennas are mounted.

The azimuth arm permits the Gregorian

dome or carriage house to be positioned

anywhere up to 20° from the vertical.

Twenty-six electric motors control the

positioning of the azimuth arm, gregorian

dome, and carriage house with millimeter

precision.

The Arecibo site was selected because of

its proximity to the equator, which

provided the ability to not only observe

the ionosphere but nearby planets as well.

The Karst terrain, notable for its large

limestone sinkholes provided a natural

geometry for the reflector. Construction

of the Arecibo observatory was begun in

the summer of 1960. The observatory

was formally opened three years later in

November 1 963 . The current high

;rade in 1 974 along with a high frequency

ailed as part of major upgrade completed in

The location of the observatory in Puerto Rico meant that the design and construction had

to account for both seismic activity and wind loads due to hurricanes. The antenna

platform and tower system is designed to withstand sustained 200 km/h winds. A seismic

evaluation of the structure is in progress to verify that the structure is still safe following

the addition of the gregorian dome, which added significant mass to the antenna platform.

The antenna platform experienced a maximum observed motion of 5 cm during

Hurricane Georges in 1998.

The delegation was given a tour of the control room for the observatory and was able to observe an experiment in progress. The delegation was then permitted to walk out to the

antenna platform along the personnel access catwalk. The visit to the platform provided

576
the delegation with a unique perspective of the scale of the structure. The members of

delegation had the opportunity to ask specific questions about the design and construction

of the observatory and considerations for extreme loads due to wind and seismic events.

Cerrillos and Portugues Dams

The delegation traveled to the city of Ponce on the southern coast of the island to visit

two U.S. Army Corps of Engineers dam projects. Ponce, because of its location on the

Caribbean coast, and the proximity of the Bucana and Portugues Rivers is subject to flooding during heavy rainfall and has experienced a number of serious floods. The

headwaters of both rivers originate in the Central Mountains, and flow southward toward

Ponce. The dams and related channel improvements are intended primarily for flood

control. Both dams have been designed so that they can be modified to provide water

supply to surrounding localities in the future. The reservoirs will also be developed to provide recreation areas for local residents.

The tour began at the field office for the two projects. The delegation was given a

presentation on the two dams and related channel improvements by Pablo Vasquez, a

civil engineer with the U.S. Army Corps of Engineers site office located near the

Cerrillos Dam. The presentation provided an historical context for the two projects

including a discussion of the flood events that triggered the projects as well as providing a thorough overview of the projects and some of the key features of the dams, channel

improvements and drop structures.

Figure 3: View of Cerrillos Dam embankment.

577
Following the presentation, Mr. John Conway of the U.S. Army Corps of Engineers took

the delegation on a tour of the Cerrillos Dam. The dam consists of a 98.5 m high zoned

rockfiU embankment with a central clay core flanked by a two-stage filter and supported

by zoned rock shells. The axis of the embankment is approximately 474 m long. A 60 m

deep, triple line grout curtain was placed beneath the dam along the centerline axis. The

crest of the dam is 193.5 m NGVD (National Geodetic Vertical Datun). The minimum

flood control pool elevation is 175 m NGVD. An emergency spillway, cut into the right

embankment of the dam has a crest of 186.3 m NGVD and is intended to prevent the pool

from cresting the main embankment. The outlet works is located in the left abutment of

the dam and includes an inclined intake structure, a 5.5 m diameter regulatory outlet

tunnel, and a stilling basin. Because the dam is subject to earthquakes, a series of strong

motion detectors has been installed at intervals across the dam.

Figure 4: Reservoir created by Cerrillos Dam.

After touring the Cerrillos Dam, the delegation was taken to view the Portugues dam

construction site. Construction of the dam itself is scheduled to begin in late 2000. At

the site, the rails and moveable platform used to carry equipment for the injection of the

grout curtain and the injection pipes were visible. Similar to the Cerrillos dam, a 60 m

deep, triple wall grout curtain was also placed for the Portugues Dam.

The Portugues Dam, when complete, will be the first of its kind constructed by the U.S.

Army Corps of Engineers. It is a three-centered concrete dam, curved in both the vertical

and horizontal directions. The dam will be constructed in two phases, the first to provide

578
flood control and recreation

and the second to provide

water supply. When

complete, the dam will have

a length of 459 m and a

crest height of 82.5 m. The

dam will be 1 3 m wide at the base and 3.7 m wide at

its crest.

The delegation was also

able to view some of the

channel improvements

made to both the Bucana

and Portugues Rivers.

Along the Bucana, it was

possible to see sections

where rip-rap had been

placed to prevent channel

erosion during periods of

high flow velocities.

During the drive to the

Portugues Dam, drop

structures and concrete

charmel walls that had been

constructed along the

Portugues River were

visible. From the top of the

Cerillos Dam, the

realignment of the

Portugues River, which

Figure 5: West side of Portugues Dam site. Grout pipes are changed the location of its

visible to right of rails. confluence with the Bucana

River, could be seen.

2. Tren Urbano

On May 23, the delegation toured the alignment of the Tren Urbano urban rail system in

the San Juan area. Approximately one-third of Puerto Rico's 3.8 million residents live in

the San Juan Metropolitan Area (SJMA), a region of about 1000 km^. SJMA residents

make approximately 3.2 million trips per day, and an estimated 1625 vehicles per square

kilometer in the central SJMA combine to produce some of the worst traffic congestion in

the world.

The initial phase of the Tren Urbano project will serve the municipalities of San Juan,

Bayamon, and Guaynabo. The line will operate 20 hours a day and is expected to serve

579
1 15,000 passengers per day. The population density within 0.8 km of the alignment

ranges from (4,000 to 8,000)/km^ and over 30 % of the regional employment will be

located within 0.5 km of the train corridor. The Phase I aUgnment is 17.2 km in length

and will have .16 stations. Fifty-two percent of the alignment is elevated, 40 % at grade,

and the remaining 8 % underground. The Phase I alignment is scheduled to open in the

summer of 2001 at a cost of $ 1 .67 billion.

The delegation began its tour at the Tren Urbano office in Hato Rey. Dr. Dru Desai, of

Daniel, Mann, Mendenhall & Johnson, gave an overview presentation of the project,

including a discussion of the project requirements, contracting, and construction

management. A second presentation oriented the delegation to the alignment and

included a videotaped flyover of the Phase 1 alignment.

Following the presentations, the delegation was taken on a tour of the alignment.

The tour began at the western terminus of the Phase 1 alignment in Bayamon. The

delegation was shown the casting yard where precast concrete segments for the elevated

guideway are produced. It was also possible for the delegation to observe the

construction of the elevated guideway. The segments are placed using an overhead

launching truss and post-tensioned with steel cables. From Bayamon the delegation was

taken to Martinez Nadal to view a station under construction, the maintenance yard, and a

section of at-grade rail that will serve as the

test track for the system.

From Martinez Nadal, the delegation was

taken to the Tren Urbano office in Rio

Piedras where it was met by Dr. Rafael

Jimenez, Project Director for Tren Urbano.

Dr. Jimenez gave a presentation detailing

the procurement process and quality

assurance/quality control aspects of the

project. Following the presentation, the

delegation was taken to see the construction

of the tunnels and station in Rio Piedras.

Underground construction was necessary in

Rio Piedras due to the presence of a large

number of historic buildings and the density

of building in the area. Two types of tunnel

construction were seen. The Rio Piedras

station is being constructed using a stacked

drift method, where a series of small turmels

are excavated and filled with concrete.

Each tunnel is offset slightly from the

tunnel immediately below it thus forming

Figure 6: Elevated guideways under an arched vault when complete. The arched

construction in Bayamon.

580
Figure 7: Station under construction in Martinez Nadai.

vault will house the Rio Piedras Station. The delegation was also shown two tunnels

formed using the New Austrian Tunneling Method (NATM). These tunnels were bored

using a tunneling machine and then lined with

precast concrete segments that are gasketed and

bolted together and held in place by earth

pressure. This construction method provides a

level of flexibility to the tunnel enabling it to withstand seismic forces.

Each member of the delegation also received

information on the design of the structures used on the project, the organization of the Tren

Urbano office, the procurement process used, and

general information about the Tren Urbano

system. This is an impressive project and will be

a world-class urban transit system when

completed.

Figure 8: Tunnel under construction in Rio Piedras. Precast concrete liner segments are visible.

AU.S. GOVERNMENT PRINTING OFF1CE:2001 -472-068/46007 5 g 1

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