rU INST OF STANDARDS & TEC^

NBS BUILDING SCIENCE SERIES

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The Building Science Series

The Building Science Series disseminates technical information developed at the National Bureau of Standards on

building materials, components, systems, and whole structures. The Series presents research results, test methods, and performance criteria related to the structural and environmental functions and the durability and safety characteristics of building elements and systems. These publications, similar in style and content to the NBS Building Materials and Structures Reports (1938-59),

are directed toward the manufacturing, design, construction, and research segments of the building industry, standards organizations, and officials responsible for building codes. The material for this Series originates principally in the Center for Building Technology of the NBS Institute

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please contact the Program Planning and Liaison Unit, Center for Building Technology, Institute for Applied

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[See mailing list announcement on last page]
] 2 3 1877 ^"^'^ Investigation of the Skyline Plaza Collapse

lo^s in Fairfax County, Virginia

Edgar V. Leyendecker and S. George Fattal

Center for Building Technology

Institute for Applied Technology

National Bureau of Standards

Washington, D.C. 20234

Sponsored by the Occupational Safety and Health Administration

U. S. Department of Labor

Washington, D.C. 20001

U.S. DEPARTMENT OF COMMERCE

Dr. Betsy Ancker-Johnson, Assistani Secrefary for Science and Technology

NATIONAL BUREAU OF STANDARDS, Ernest Ambler, Acting Director

Issued February 1977
Library of Congress Catalog Card Number: 77-600002

National Bureau of Standards Building Science Series 94

Nat. Bur. Stand. (U.S.), Bldg. Sci. Ser. 94,91 pages (Feb. 1977) CODEN; BSSNBV

Supersedes NBSIR 73-222

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 1977

For sale by the Superintendent of Documents, U.S. Government Printing Office, Washington, D.C. 20402

(Order by SD Catalog No. CI3.29/2:94). Stock No. 003-003-01696-9 Price $1.55

(Add 25 percent additional for other than U.S. mailing).
SI CONVERSION UNITS

In view of the present accepted practice in this country for building

technology, common US units of measurement have been used throughout this

paper. In recognition of the position of the United States as a signatory to the General Conference on Weights and Measures, which gave official status to the metric SI system of units in 1960, assistance is given to the reader

interested in making use of the coherent system of SI units by giving conversion factors applicable to US units used in this paper.

Length

1 in = 0.0254 meter (exactly)

Force

1 lb (1 bf ) = 4.448 Newton (N)

Pressure

1 psi = 6895 N/m^

Tempera ture

5/9 (Temperature °F-32) = Temperature °C

iii

TABLE OF CONTENTS

Page

1 . Introducti on 1

1 . 1 Background 1 1.2 Object iveandScope. . 2

2. Description of the Structures and Their Collapse 5

2. 1 Introduction 5 2.2 Building A-4 5 2.2.1 General 5 2.2.2 Formwork and Interior Views 18

2.2.3 Erection Cranes 31 2.3 Parking Garage 45

2.4 Crane No. 2. 51

3. Concrete and Reinforcement in Building A-4 57

3.1 Concrete 57

3.1.1 Standard Cylinder Tests 57

3.1.2 Field Cores 57

3.1.3 Estimate of Quality of Concrete as Delivered 57

3.1.4 Estimates of Concrete Strengths on the Day of

Col 1 apse 59

3.2 Reinforcing Steel 62

4. Structural Investigation of Failure Conditions in Building A-4. . 65

4.1 Introduction 65

4.2 Controlling Regulations and Basis of Design 65

4.3 Analytical Procedure 65

4.4 DiscussionofResults 68

4.5 Probable Mode of Collapse in Building A-4 32

5. Summary ofFindings 85

5.1 Mode of Failure 85

5.2 Non-Compliance with OSHA Regulations 85

6. References 87

V

INVESTIGATION OF THE SKYLINE PLAZA COLLAPSE IN

FAIRFAX COUNTY, VIRGINIA

Edgar V. Leyendecker and S. George Fattal

The collapse of the Skyline Plaza apartment building A-4

has been studied by using information contained in case records

of the Occupational Safety and Health Administration (OSHA),

U.S. Department of Labor and obtained from on-site inspections

by investigators from the National Bureau of Standards.

Non-compliance with OSHA construction standards has been

identified with regard to formworkj field-cured concrete specimens

and crane installation. Specifically , the construction procedures

did not comply with standards for the removal of supporting forms.

It is concluded that premature removal of forms was a contributing

factor to the collapse in building A-4.

An analysis of the 23rd-floor slab indicates that its most

likely mode of failure was in shear around one or more columns

in section 3 of the floor slab. The strength of the 23rd-floor

slab on the day of collapse has been estimated to be at a level

that removal of shoring could have produced shear failure in the

slab.

Key Words: Apartment building^ collapse; concrete; concrete

strength; construction; flexure; progressive collapse;

shear; strength.

CHAPTER 1 INTRODUCTION

1 . 1 Background

The Skyline Center Complex located near Bailey's Crossroads, Fairfax

County, Virginia is a development planned to contain eight apartment buildings, six office buildings, a hotel, and a shopping center [19]. Two

apartment buildings which have been completed are shown in figure 1.1. A pair of apartment buildings (similar in appearance to those in figure 1.1)

and an adjoining parking and lobby structure were under construction and

included the structures which collapsed on March 2, 1973.

*Dr. Edgar V. Leyendecker and Dr. S. George Fattal are structural research

engineers with the Structures Section; Structures, Materials and Life

Safety Division; Center for Building Technology; Institute for Applied

Technology; National Bureau of Standards; Washington, D.C. 20234.

^Numbers in brackets refer to references.

1
The apartment buildings under construction are shown in figure 1.2, an aerial photograph taken at about 11:00 a.m. on Friday, March 2, 1973, from an altitude of 5000 ft [18]. Several hours later, at about 2:30 p.m., a portion

of the building shown in the top of figure 1.2 collapsed. The collapsed

portion of the building was located approximately under the slab area being

cast (in figure 1.2) and extended vertically for the full height of the

building, 23 stories plus four basement stories. The collapse progressed

horizontally from the tall building to include the entire parking garage area

and stopped at the building shown at the bottom of figure 1.2. The full

extent of the collapse is shown in figure 1.3. It has been reported by OSHA

that fourteen construction workers were known to have been killed, four in

the garage and ten in the tower, and another 34 injured in the incident.

Within a few hours of the incident an inspection team from the Occupational Safety and Health Administration (OSHA), Department of Labor, began

arriving at the site to begin an investigation into the collapse.

1 . 2 Objective and Scope

On Monday, March 5, OSHA requested the technical assistance of the Center

for Building Technology of the National Bureau of Standards (NBS) with respect

to the collapse. The National Bureau of Standards was requested to ascertain,

if possible, the cause of the incident, to assist OSHA in determining whether

there had been non-compliance with OSHA standards ("Safety and Health Regulations for Construction" [lOj) and whether such non-compliance contributed to

the collapse.

OSHA compliance officers were on the site from March 2 through March 16

collecting material for their case records [18]. During this time, personnel

from the NBS made numerous site inspections.

The NBS investigators used data gathered during on-site inspections,

OSHA case records, structural and architectural drawings, shop drawings, and

structural computations in preparing this report. Where reference is made

in this report to employee statements, such statements are part of OSHA case records examined by NBS.

2
,1

I

Figure 1.1 Appearance of completed apartment buildings .

I

Figure 1.2 Aerial view of construction at 11:00 a.m.,

March 2, 197Z (north is to the top).

3
4
CHAPTER 2 DESCRIPTION OF THE STRUCTURES AND THEIR COLLAPSE

2 . 1 Introducti on

Three structures may be identified in figure 1.2; building A-4 at the

top, building A-5 partially shown at the bottom, and a parking garage in

between the two buildings. The collapse started in building A-4 and progressed

vertically to the ground and horizontally to include the entire parking garage.

The collapse stopped at building A-5 which was structurally isolated

from the parking garage. The two affected structures are shown in a plan

view in figure 2.1 and are discussed in subsequent sections.

The three structures under construction were designed under the Fairfax

County Building Code Ordinance [5] which incorporates by reference the provisions of the American Concrete Institute's Building Code Requirements for

Reinforced Concrete (ACI 318-63) [6]. The bu i 1 d i ng des i gn is discussed

further in section 4, Structural Investigation of Failure Conditions in

Building A~4. Applicable Federal safety regulations are described in the

Safety and Health Regulations for Construction [10] which incorporates by

reference the American National Standards Institute's Safety Requirements

for Concrete Construction and Masonry Work, ANSI AlO. 9-1970 [2].

2 . 2 Building A-4

2,2.1 General

Building A-4 was of reinforced concrete flat plate construction supported

on a 4-ft thick foundation mat. The completed structure was to have 26 stories

of apartments, plus a penthouse and a four-story basement (designated B-1,

B-2, B-3, and B-4). The typical story height from the first story up was 9 ft

0 in from top of slab to top of slab. Floor slabs were 8 inches thick.

The basement story heights varied in order to suit mechanical equipment

layout with a total basement height of about 40 ft as measured from the top

of the foundation mat (level B-4) to the top of the first floor slab. The

basement floors were about the same elevation as the corresponding floors in

the garage, although there was no actual floor B-3 in building A-4. That is,

the lowest "story" in the basement consisted of stories B-4 and B-3. The

first floor slab was at the same elevation as the roof of the garage which

was to become a landscaped area.

The plan view shown in figure 2.1 is that of the 22nd story and is typical

for the 1st through 26th stories in the column layout. The column layout

remained essentially the same through the height of the building (1st story

and up) varying only in size, reinforcing steel, and concrete strength. There

are eight shear walls in the structure. These are designated as A through H in the drawing. The configuration of the shear walls above the 20th story is

shown in figure 2.1. For the 20th story and below, the portion shown dashed

became part of the shear wall. The floor slab thickness of 8 in was constant

through the height of the building. A 1/2-in expansion joint separated the

building into two parts at grid line H.

5
6
Normal weight aggregate concrete was used in the columns. The specified

column concrete strength varied; it was 5000 psi from the foundation mat to

the 7th floor, 4000 psi from the 7th to the 17th floor, and 3000 psi above

the 17th floor. The slabs used lightweight aggregate (coarse aggregate only)

concrete with a specified strength of 3000 psi. Inspection of two floors,

the 24th and 10th, indicated that the lightweight aggregate concrete floor

slab passed through the columns at these floors.

A typical floor of the building was poured in four sections; the progress

of construction at the time of the collapse is shown in figure 2.2 [18]. The

actual sizes of the pour sections are shown in figure 2.1. Note that the

sections are not equal in size and configuration. According to statements

provided to OSHA, a normal construction rate was one section a day. This

rate of construction would permit the completion of one floor per week and

allow an extra day for weather variation. In actual practice, this rate was

not always maintained as indicated in figure 2.3.

Figure 2.3 is a plot of daily temperature versus calendar date for

January 28 through March 2, the day of the collapse. The maximum, minimum,

and average temperatures are those recorded at Washington National Airport,

the official weather bureau recording station nearest the construction site.

Where available, the daily temperature range recorded at the construction

site (from job records) is also plotted. The dates for pouring the various

sections of the 21st through 24th floors are indicated on the drawing. Note

that the casting dates shown tended to fall on the days with the higher

temperatures

.

The general appearance of the building A-4 as viewed from the southeast

after the collapse is shown in figure 2.4. A number of the floors and one

of the columns is identified for later reference. Note the absence of a floor slab at level B-3 (in accordance with the plans). Note also the column

corbels at floor levels B-3, B-2, B-1 , and 1. These corbels supported one edge of the parking garage slabs. Level B-4 of the garage was on grade.

A partial view of the north face of building A-4 is shown in figure 2.5.

A number of columns and floors are identified. The collapse extended between

shear wall H and column 33 on the south face, a distance of about 65 ft (refer

to figure 2.1), On the north face the collapse extended between columns 12

and 17, a distance of about 104 ft. Note that for most floors the slab

between columns 16 and 17 did not sever from the main building but is sagging

for a one-story height. Below the 20th floor, the horizontal extent of the

collapse is as discussed above. Above the 20th floor, the failure zone extended slightly to the west and a greater amount to the east.

Figures 2.6 through 2.8 are closeup views of the east end of the failure

zone. These figures require little comment at this time; however, it should

be noted that the formwork is clearly visible in figure 2.6.

A general view of the west end of the failure zone is shown in figure

2.9 along with closeup views in figures 2.10 through 2.12. These figures

also require little comment at this time; however, it should be noted that

the formwork is clearly visible in figure 2.10.

On the night of March 4 the remaining portion of the building to the

east of the failure zone was completely demolished, with the result shown

in figures 2,13 and 2,14,
Not poured- Not poured

Columns being poured

Poured 2/28/73-^ Poured 3/1/73-^

Poured 3/2/73,

finishing in progress-^

Not poured^

Poured 2/23/73^ Poured 2/24/73-^ Poured 2/26/73-^ Poured 2/27/73-^

Poured 2/14/73^ Poured 2/15/73-^ Poured 2/20/73-^ Poured 2/21/73-^

Poured 2/5/73-^ Poured 2/6/73^ Poured 2/7/73-^ Poured 2/13/73-^

Section 1 Section 2 Section 3 Section 4

25th floor

1 24th floor

23rd floor

22nd floor

21st floor

View is looking north Figure 2.2 Extent of construction at the time of collapse.

Figure 2. 3 Temperature and casting history for 2lst story and up.

COLUMNS

FLOORS

CALENDAR DATE

8
Figure 2,4 General appearance of building A-4 as

viewed from the southeast

.

9
Figure 2.5 General view of the north face of building A-4

.

10

Figm>e 2 . 7 Closeup view near the

top of the east end of

the failuve zone

3

looking south.

Figure 2 . 8

Closeup view near the

midheight of the east

end of the failure

zone, looking south.

12
Figure 2.9 General view of building A-4 as seen looking

southwes

t

.

13
Figure 2.10 Closeup view of the top of the west end of the

failure zone ^ looking southwest

.

14
Figure 2.11

Ctoseup view near the

midheight of the west

end of the failure

zone 3 looking southwest.

Figure 2.12

Closewp view at the

bottom of the west end

of the failure zone

looking southwest.

15
Figure 2.13 Appeavanoe of A~4 after the removal of the east

end of the building ^ viewed from the east.

16

2.2.2 Formwork and Interior Views

Access was not gained to building A-4 by OSHA personnel until March 5 when the building appeared as in figure 2.14. Most of the photographs used

in this discussion of the building interior were taken on or after that date.

Between March 2 and March 5, additional shoring was placed in some parts of

the building.

Figure 2.15 shows the locations of the formwork on March 2, shortly after

the collapse. Full formwork was in place on the 24th and 23rd stories in

sections 1 and 2 (refer to figure 2.1 for location). Some formwork may be seen on the 22nd story in sections 2 and 3. In section 2 this is primarily

the area around the material elevator shown toward the western end of the

south elevation in figure 2.15. Investigation by OSHA personnel on March 5 indicates that the only formwork not stripped on the 22nd story in section 2 was around that elevator. Employee statements indicate that this area was

not stripped in order to prevent lumber from falling on bricklayers working

below. Formwork may also be seen in section 3 of the 22nd story in figure

2.15. However, closer examination of the 22nd story in figure 2.10 indicates

that formwork was stripped in the center portion of the 22nd story.

Full formwork was also in place under section 4 on the 22nd and 23rd

stories. No reshoring can be seen under section 4 below the 22nd floor in

figure 2.15 or in figures 2.4, 2.5, and 2.6. Some employee statements indicate,

however, that reshoring was present below pour 4 on the 21st story. Other

statements contradict this.

There is conflict also among employee statements as to the formwork which

was in place in section 3. Employees interviewed by OSHA agreed that the

23rd-story forms supporting the recently cast 24th floor were in place. However, employee statements indicate that the 22nd-story forms were: (1) entirely

removed, (2) partially removed, or (3) not removed. The same contradictions

were indicated for reshoring on the 21st story. The angle of the aerial

photograph in figure 1.1 prevents examination of conditions of formwork in

section 3. The location of formwork will be discussed further after a description of the interior of the building.

The formwork on the 24th story is shown in figure 2.16. Several columns

have been identified for reference (see figure 2.1 for location). These slab

forms had not been completely erected. A schematic of this formwork is

shown in figure 2.17 as derived from OSHA case records [18].

The formwork in figure 2.17 is typical for the floor slabs with design

noted by the concrete contractor as being based on the publication,

Formwork for Concrete , SP-4 [11]. Formwork sheets submitted to Fairfax County

[18] contained no mention of lateral bracing for the form system. However,

a limited amount of bracing may be seen in figure 2.16 and some subsequent

figures. OSHA regulations (ANSI-A10.9, Sections 6.3.2 and 8.1.5) [2]^ require

a lateral bracing system capable of resisting a lateral force of at least 2 percent of the dead load of the slab.

"^For relationship between OSHA regulations and ANSI-A10.9 see section 2.1.

18
Figure 2.15 General appearance of building A-4 as viewed

from the southeast

.

19
20
As may be seen in figure 2.16, the stringers were placed east-west and

the joists were placed north-south. Employee statements indicate that stringers

and joists were usually 16 ft in length although some were as short as 6 to

8 ft.

The erection procedure is .described by reference to figure 2.16. Stringers

were erected with a shore under each end (shores 1 and 5). These shores were attached to the stringers by a metal or wooden plate. Shores 2, 3, and 4 were

then inserted in basket-like sockets which were fastened to the stringers.

Employee statements indicate that diagonal bracing was installed in the

north-south direction on every 4th shore at approximately 16-ft intervals

(also see figure 2.20). The joists were placed over the stringers at about

16-in intervals (figure 2.17 indicates the spacing could be 20 in). Joists

were toenailed to the stringers at about 8-ft intervals. Plywood sheathing

was then placed on the joists with the long dimension parallel to the stringers.

Forms were removed by knocking out shores 1, 2, 4, and 5 in figure 2.16.

Shore no. 3 would then be knocked down which would allow the stringer to fall.

The joists and plywood would then be pulled down.

Usually, the center portion of the slab would be stripped first. Approximately 10 stringers would be pulled down before reshoring began. The area stripped prior to reshoring was approximately 20 ft wide by 35 ft long.

During the NBS site inspection, a considerable portion of the lumber

used for the remaining forms in the 24th story was found to be in poor

condition or out of plumb. For instance, the top photograph in figure 2.18

shows a stringer with a battered end directly over the shore. The second

photograph illustrates a shore which is out of plumb and has a vertical crack

near the top. OSHA regulations (ANSI-A10.9, sections 8.1.24 and 8.1.25) [2]

require correction of both of these conditions prior to placing concrete (which

was not yet placed on these forms).

The portion of section 3 of the 24th story slab which remained standing

is shown in figure 2.19. This portion of the slab was several hours old at

the time of the collapse. Apparently, the only activity on the slab at the

time of collapse was work being done by concrete finishers. One concrete

finisher working near the stair well (around columns 65 and 66) indicated

that a large deflection was seen in the middle of section 3. The statements

of numerous employees agree on this point. No statements provided an exact

location of the sag. However, the middle of section 3 is about midway between

columns 67 and 85 (refer to figure 2.1 for location).

The formwork on the 23rd story is shown in figure 2.20. These forms

supported the 24th floor which had been cast on February 28 (section 1) and

March 1 (section 2). Note the number of shores out of plumb and the location

of lateral bracing. The method of attaching the brace to the stringer may be seen in figure 2.20. The method of attaching the brace to the floor is shown

in figure 2.21. A portion of a nominal 3x4 was nailed to a piece of

plywood which was in turn nailed to the floor. A plywood plate was occasionally

used to nail the brace to the nominal 3x4. Note the absence of such a plate

in figure 2.21. Occasionally, the brace was installed with the plate, in other

instances, there was no positive attachment of the brace to the floor. In one case, the plywood stop was not nailed to the floor, providing no lateral

resi stance

.

21
,3"x4" JOIST at 20" ON CENTERS

5/8" PLYWOOD

5'- 9" U

-3"x6" STRINGERS

_3"x 4" SHORES ( 7'-6-3/4") OR

ELLIS CLAMPS

5-9

Lumber- I450fand better Plywood- Douglas fir, Class I, B/B exterior 2000 f Figure 2,17 Formwork for a typical floor.

Figure 2.18 Selected details of 24th story forms.

22
23
24
25
As in the case of the 24th story, the 23rd story lumber which was apparently damaged was reused. Examples of battered shores are shown in figure

2.22. Note the large reduction in bearing area. As in the case of the 24th

story, OSHA regulations (ANSI-A10.9, sections 8.1.24 and 8,1.25) [2] required

correction of these conditions prior to placing concrete.

Reshoring in section 2 of the 22nd story as of March 6 is shown in figure

2.23. OSHA investigators have indicated that the reshoring on March 5 was

not as extensive as shown in figure 2.23. Apparently on March 5 the reshores

were principally in place on the balconies, as they were on March 3, the day

of the collapse.

Careful examination of figure 2.23 reveals that the center ceiling area

of section 3 (area beyond column 59) was bare, indicating that some stripping

had occurred in section 3 (see also figure 2.10). Although there is conflict

in their statements, a number of workmen that escaped the building by way of

the stairs in section 4 (see figure 2.1) have indicated that the 22nd story

was either partially or entirely stripped in section 3. Notes on the engineer's structural drawings call for two full stories of shoring and one story

of reshoring under a slab being cast.

A view of a portion of section 2 of the 21st story is shown in figure

2.24. Column 59 is at the expansion joint (Grid H in figure 2.1). Note the

cables on the floor which were installed after the collapse to anchor to

section 2 the portion of section 3 which was still standing. No reshoring

was present on this floor in sections 1 or 2.

A number of workmen have stated that at least some reshoring was present

in sections 3 and 4 of the 21st story. The statement of one employee indicates

that he was working in section 3 installing reshores. At about 2:00 p.m. he

could hear the stripping crew working on the next floor up (22nd floor). He

could hear the noise caused by the falling lumber. At about this same time,

the workman's reshores began falling so that just before the accident there were no reshores in section 3 (this point is discussed later).

As shown in figure 2.25, the 20th floor was completely bare of reshoring.

This was reportedly the case for sections 1 through 4. Note the presence

of shear walls A and B. These walls were not present (the design did not

require them) above the 20th story. Note also the erection crane which is

barely vi si bi e to the left of the designation for shear wall A. A view of

the failure section at the end of the 20th floor is shown in figure 2.26.

It has been indicated in this report that there is conflict in the

various employees' statements as to what formwork was in place at the time

of the incident. However, it should be noted that conditions changed throughout the day as construction proceeded; that is, what may have existed at one time may not have existed a short while later. Nevertheless, it is necessary

to determine the location of forms and reshores at the time of the incident.

The estimated location of forms and reshores is given in figure 2.27

and discussed below. Physical evidence examined by OSHA and NBS investigators

as employees' statements indicate that all formwork was in place for all

poured sections on the 23rd and 24th stories. Based on employees' statements

and careful observation of photographs taken on the day of the collapse and

Skyline Center Structural Drawings, Weihe, Black and Jeffries, Architects,

dated March 16, 1972, with revisions through October 2, 1972 - Drawing 55.

26
Figure 2.22 Selected details of 22rd story forms.

27
28
Figure 2.24 Section 2 of 21st story , looking east.

I

I

i

Figure 2.25 Section 1 and 2 of 20th story y looking east.

29

I

I

I

i
Figure 2.26 Seation 3 of 20th story, looking east.

Figure 2.27 Estimated location of forms and reshores

at the time of the collapse

.

^'^ured 3/1/73^^^^^ R3ured 3/2/73^ indicates not poured^

^-^^ured 2/23/73V^\ ^-^oured 2/24/73^^^\ ^^Poured 2/26/73^^^\

indicates formwork in piace^^p<:

^^oured 2/27/73^\^

Poured 2/14/73^ Poured 2/15/73^

stripping in' progress

Poured 2/20/73-^^-^ ^H'oured 2/21/73^^\^

1 1 some,ifany, reshoring

Poured 2/5/73^ Poured 2/6/73^ Poured 2/7/73^ Poured 2/13/73--^

SECTION 1 SECTION 2 SECTION 3 SECTION 4

25th Floor

24th Story

24th Floor

23rd Story

23rd Floor

22 nd Story

22 nd Floor

21st Story

2lst Floor

LOOKING NORTH

30
over the following days, it is concluded that all formwork had been removed

from sections 1 and 2 of the 22nd story on the day of collapse. The one exception is in the limited area near the materials elevator on the south

face of the building. It appears that very little reshoring had been done.

Examination of photographs indicates that formwork was in place in section

4 of the 22nd story. Section, 3 is discussed later. On the 21st story there

was no reshoring in sections 1 and 2. One statement indicates that the re- shoring had been removed from section 2 on the day of the accident. Although

some statements indicate the presence of reshores in section 4 of the 21st

floor, this is not supported by photographic evidence available (figures 2.4,

2.5, 2.6, and 2.15). The photographs at the angle at which they were taken

would not show reshoring, if any, located at some distance away from the

edge of the slab. It is considered possible that there was some reshoring

present although probably a small amount. Section 3 of the 21st story is

discussed below.

One workman indicated that, at the time of the incident, he was placing

reshores in section 3 of the 21st story and that some reshores were present

when he started work. Prior to the incident all of the reshores fell out

(except those in the balcony areas). This is consistent with what could occur if the forms had been removed in the story above. For example, consider

figure 2.28 (a) which shows a frame with forms in the 3rd and 4th stories and

reshores in the 2nd story. The fifth floor has been freshly poured and carries

none of its own weight which must be distributed to the 4th, 3rd, and 2nd

floors. The exact distribution of loads depends on a number of factors, one

of which is the construction history.

If the forms are removed in the 3rd story as in figure 2.28 (b), then

the 4th floor carries all of its own weight in addition to that of the fifth

floor. The 2nd and 3rd floors are now relieved of the previous load from

the 4th and 5th floors. The 3rd floor deflection will decrease due to this

reduction of load. A possible smaller contribution to the decrease in the

3rd floor deflection is due to the downward deflection of the 4th floor under

the added load causing the third story column to bend outward and the 3rd

floor to bend upward as schematically illustrated in figure 2.28 (c). The

net result of the removal of forms causing an upward deflection of the 3rd

floor could quite conceivably cause the 2nd story reshores to fall out with

the final result shown in figure 2.28 (d).

Based on the above simplified analogy plus previously mentioned photographs (figures 2.10, 2.15, 2.23) it is concluded that form stripping was

in progress in the 3rd section of the 22nd story. The photographic evidence

indicates that at least the central portion of the section was being stripped

heading toward section 4.

2.2.3 Erection Cranes

Two climbing cranes were used in the construction of building A-4.

Crane no. 1 was located in section 2 and crane no. 2 was located in section 4. The terminology used in the discussion of the cranes is shown in figure

2.29 [22].

Both cranes were initially used in a free-standing position mounted to

the foundation mat. Later, as construction proceeded, the cranes were supported by floors within the building. Except for certain installation

requirements, crane dimensions, and capacity, the two cranes were quite

similar. Descriptive data are contained in figures 2.30 [21] and 2.31 [22]

for the two cranes.

31
/Fioor no

^ ^Freshly poured -Beam

Forms

Reshores

a) forms and reshores

in place

-Column

b) third-story forms

removed

c) deflected shape

of third story

5

3

2

d)final results

Figure 2.28 Effect of removal of forms.

Figure 2.29 Crane terminology

\ \ \ \ \ \

« 29 31 \

FREE- STANDING CRANE

TOVER BASIL PLATE BASC TOWER SECTION STANDARD TOnCR SCCTiOH

SLiPRiNG TOMER SECT ION RiNGOEAR ASSEMSLt

JtB PIVOT FRAHC TOWER TOP SECTION

JIB BUTT SECTION (i) JIB SeCTIOK (21 JIB SECTION ( SI JtB SECTION (4) JIB SECTION (9)

JIB SECTION («)

v'lS SCCTtON (7) counter jib butt section counterjib section id counterjib section IZ) Jta SUSPENSION ROO

JIB SUSPENSION ROD (EXT ) JIB Suspension rod (ex.t)

Jte SUSPENSION ROD lExT } COUNTERJIB SUSPENSION ROO

COUNTERJIB SUSPENSION ROD lExT ) OPERATORS CABIN

TROLLf

T

LOAD BLOCK

HOIST MACHINERT

TROLLEY HOTOR 8 WINCH HOIST CABLE

TROLLEY CABLE COUNTER WEIGHT

HOIST CONTROL CABINET MOMENT OVERLOAD DEVICE TROLLEY STOP

SLIPRtNe

CLIMBING MACHINERY

CLIMBING SUPPORT BEAM

CLIMBING PrtAME CATWALK AND HANDRAIL TOWERTOP LADDER TOWER LADDER

SWING-a TROLLEY CONTROL CABINET

CLMBINC LADDER

BOGIE UNDERCARMIACE FRAME

OUTEA TOWER BASE SECTION OUTER TOWEH STANDARD SECT. ON

KNEE BRACE

CONCRETE BALLAST BLOCKS

TRAVEL MOTOR ASS M

?! TRAVEL CONTROL CABINET

S4 POWER CABLE DRUM

TRAVEDNG UNDERCARRIAGE WITH OUTER TOWER

32
SPECIFICATIONS ARE

SUBJECT TO CHANGE

WITHOUT NOTICE SLAB OPENING

COWFlGORATlQN B/1

ruT mom ( o* tovi* - LOADtNG DIAGRAM

Figure 2.30 Crane no, 1 speaifiaations.

Figure 2.31 Crane no. 2 specifications

33
The standard climbing crane consists of the tower base section (which

contains the climbing machinery, no. 2 in figure 2.29), four standard tower

sections Cno. 3 in figure 2.29), and a tower slipring section (no. 4 in

figure 2.29). The four standard tower sections are equal in length and may

be interchanged. The ring gear assembly (no. 5 in figure 2.29) which forms

the connection between the stationary tower and the rotating portion of the crane fits on top of the slip ring section.

A schematic of the climbing sequence is shown in figure 2.32 [22]. In picture 1 the crane is installed in its initial position on the foundation

mat. A lower support, the climbing frame, is installed a maximum of 36 ft

from the mat. The climbing ladder is suspended from the climbing frame (the

ladder is about 38 ft long). An upper support, consisting of four corner

clamps, is attached a specified distance above the lower support. This re- quired distance was specified as 21 '-4" for crane no. 2 (figure 2.31) and

18'-4" for crane no. 1 (figure 2.30). In the actual use, a two-story distance

of 18'-0" was used for crane no. 1 and apparently for crane no. 2. In picture

2, the tower climbs up the ladder to the first support using the climbing

mechanism included in the tower base section. In picture 3 a third support

is added, the climbing ladder is moved up to a second climbing frame and the crane is ready for climbing. In the fourth picture the crane is in position

and the lower frame can be removed.

The upper support transfers only horizontal forces to the building, while

the lower support transfers vertical and horizontal forces. The maximum free

standing height of the crane above the upper support as measured from the

bottom of the jib is 70 ft 2 in for crane no. 1 (figure 2.30) and 81 ft for crane no. 2 (figure 2.31).

Crane no. 1 was supported by the climbing frame,' the lower support, on the 20th floor as shown in figure 2.33. The upper support was provided on the 22nd floor as shown in figure 2.34. This total distance is 18 ft, slightly

less than the manufacturer's specified 18 ft 4 in. A typical corner clamp

detail on the 22nd floor is shown in figure 2.35. Note the cracked slab

where the clamp is attached.

Crane no. 2 was reported to be on either the 14th or 17th floor with

most reports indicating the 17th. Reported conditions of shoring under both cranes varied from no shores to shores all the way to the foundation mat.

Crane no. 1 has been positively located with its base support on the 20th

floor. Shoring under that crane is less certain. OSHA investigators have

indicated that it was not shored to the foundation until after the collapse.

In order to determine more accurately the location of crane no. 2 it

is necessary to look at previous construction photos. Such photos were available from the OSHA file [18]. However, the exact dates of the photos

are not known. It is known that construction records indicate that crane no.

1 was last raised on February 26 and crane no. 2 on February 7. Figure 2.36 contains two construction photos reported as taken around

February 10. This date is in error as will be shown. Since the lower floors

are not visible it is not possible to tell from figure 2.36 where the cranes

are located. Figure 2.37 shows a photograph taken on March 6 that can be

used to establish the location of the 15th floor in figure 2.36. A certain

pattei"' of "color" may be seen in the doors and door frames at the elevator

lane js. A comparison of the patterns between figures 2.36 and 2.37 (taken

Mar 5) establishes the location of the 15th floor in figure 2.36 (It will

be seen later that the same door pattern existed in a set of photos taken

between figure 2.36 and 2.37).

34
Figure 2.33 Climbing frame for arane no. 1

35
Figure 2.34 Upper support for arane no. 1.

Figure 2.35 Corner alamp detail of upper support.

36
37
Figure 2.37 Appearance of building A-4 on March 6.

38
Examination of figure 2.38 establishes the lower crane support on the

10th floor and establishes that the crane, as installed, could operate through

a ten-story height of building. It should be noted that the crane had marginal

clearance to cast the 20th floor although it appears that it could. The photos

were not clear enough to determine shoring conditions under the crane. Close examination of figure 2.38 shows openings in section 4 of the 20th

floor forms. Therefore, the 20th floor had not been cast at the time of the

photo. Reference to figure 2.3 indicates that this section was cast on February 1. Construction records indicate that framing for the 20th floor

forms, which is visible in figure 2.38, was installed on January 27. Based

on this evidence it appears that figures 2.36 and 2.38 were taken between

January 27 and February 1. Therefore the photograph was taken prior to February 7 which was the last date crane no. 2 was raised.

Figures 2.39 through 2.42 were reported taken on February 11 or February

18, both dates after crane no. 2 was raised. The floor numbers can be

established simply by counting at the east end of the building. Note the

location of the 15th floor with respect to the top elevator landing. This

location is in agreement with photos taken before and after figure 2.39. The

elevator door pattern can be clearly seen in figure 2.40. Note also that at

the time of the photograph, crane no. 1 is based on the 16th floor (crane no.

1 was raised on February 26) and crane no. 2 is based on the 14th floor.

Since crane no. 2 could work through 10 stories (figure 2.38) it could have

been used to cast section 3 of the 24th floor while based on the 14th floor.

The various tower sections of the crane are identified in figure 2.41.

Note that five standard tower sections were used instead of the usual four

(figure 2.31 and reference 22). Assuming the upper and lower supports were

two stories apart as has been stated, the crane had a free standing height

of about 99 ft which exceeds the maximum free height of 81 ft shown in figure

2.31. It is not known if any special precautions were taken to allow the

difference between the actual and the recommended installation.

Examination of figures 2.40 and 2.42 indicates that either the crane shoring or the formwork for the concrete closure slab of the crane opening

was in place in the third through thirteenth stories on the day of the

photograph. Conditions below the third story are not visible in the photograph.

The date of the photograph (reported as February 11 or 18) can be checked

by looking at figures 2.39, 2.42, and 2.2. Close examination indicates that

none of the 22nd floor had been cast at the time of the photograph (openings

can be seen in the slab forms). Since figure 2.2 indicates that section 1 on the 22nd floor was cast on February 14, it is established that the photograph was taken before that date. Due to the construction progress shown in

the figure it can be stated that section 2 of the 21st floor which was cast

February 6 (figure 2.2) was in place. Based on this evidence it appears that

the photograph was taken between February 6 and February 14.

Although the forms are in place for sections 3 and 4 of the 21st floor,

it cannot be positively stated that they have been cast. Therefore, on the

basis of these photographs (figures 2.39 through 2.42) it cannot be conclusively stated that the crane which was raised on February 7 was based on the

14th floor.

39
40


43
44
However, based on figures 2.36 and 2.38 the crane no. 2 was attached to the 10th floor between January 26 and February 1. Since this crane was raised

only once after February 1, it must have been located on the 14th floor at

the time of the collapse.

It has been established by physical examination that crane no. 1 was based on the 20th floor. Examination of photographs and contractor's records

have been used to conclude that crane no. 2 was based on the 14th floor.

Crane no. 2 was located in the failure zone. The position of the crane after the collapse will be discussed after consideration of the parking

garage

.

2 . 3 Parking Garage

The parking garage was a flat plate structure of po s t- ten s i one d unbonded

concrete construction. The completed structure was to have four stories of

parking with a landscaped roof. The B-4, or lowest level, was a si ab-on-grade

.

The parking garage slabs were at approximately the same elevations as the

corresponding floors in the building A-4 basement.

The plan view is shown in figure 2.1. A typical panel was 28 ft by 30

ft with an 8-in slab. The columns were supported on footings. The story

height was 9 ft for stories B-4, B-3, B-2, and varied from 10 to 14 ft for

story B-1 . Normal weight aggregate concrete with a design compressive strength

of 4,000 psi was specified throughout the structure.

Slabs B-4 and B-3 had been cast and slab B-2 had been cast to the extent

shown in figures 2.1 and 1.2. Note that in the vicinity of grids lines BB to FF and grid line 1 slab B-3 was placed on compacted fill. Compacted fill was apparently placed around the three columns, BB-1, DD-1 , and EE-1 in story

B-4. The footings for these columns were a few feet higher than the rest of

the col umn f ooti ngs .

The parking garage, as seen from building A-4 is shown in figure 2.43.

A number of columns have been identified for reference. At the time of the

collapse, slab B-2 had been cast to the extent shown. The rest of B-2 had

been formed (some of the forms had been cleared at the time of the photograph).

A closeup of several of the columns in figure 2.43 are shown in figure 2.44.

The locations of slabs B-3 and B-2 are shown in figure 2.44 on column

DD-4. Note the virtual lack of shear cones at the slab levels on most of

the columns in that figure.

The garage slabs generally came "straight" down except near the slab edges.

In the latter cases the columns usually failed in bending. Figure 2.45 shows

a view of slab B-2 with the edge between columns JJ-5 and JJ-8 visible on the

left. The view of this edge from the south is shown in figure 2.46. Note that

the columns failed at the B-3 level. Formwork can be seen lying on top of

the slab B-3. These forms apparently were in place under slab B-2.

Column KK-5 may be seen in figure 2.47. Once again the shear cone usually

associated with a slab punching failure is not visible in the photograph.

Note the ruptured reinforcing bars at the base of the column. These are shown more clearly in figure 2.48 (but not for column KK-6).

45
46
47
Figure 2.45 View looking west of stab B-2 between aolumn

lines HH and JJ.

Figure 2.46 Edge of slab B-2 at columns JJ-5 and JJ~6.

48

I
The east edge of the garage was supported by columns as shown in figure

2.49. The northern edge was simply supported by corbels on the columns.

These corbels are visible in figures 2.49 and 2.50. The pool deck level

(corresponding with the first floor of building A-4) was to rest on corbels

on columns 25, 26, 27 and 33-37. Slab B-1 was to rest on corbels on columns

25, 26, 27, 29, 31-38, and shear wall H. Slab B-2 was supported by corbels

on columns 25, 26, 27, 29, 31-38, and shear wall H. Slab B-3 was supported

by corbels on columns 25, 26, 27, 29, 31-36, and shear wall H.

The loss of columns between shear wall H and column 33 meant the loss

of two columns (31 and 32) which provided support for slabs B-2 and B-3.

The total span between column 33 and shear wall H was about 65 ft.

The west edge of the garage was supported by bearing on a wall and

framing into a ramp as shown in figure 2.51. Slab B-2 may be seen resting

on the portion of B-3 cast on a s 1 ab -on -g rade (vicinity of column EE-1).

More detail of the ramp framing can be seen in figure 2.52. The ramp was of

reinforced concrete construction. The west wall south of the ramp can be seen in figure 2.53 and with more detail in figure 2.54. The formwork and

strand which had been positioned for slab B-2 can be seen in figure 2,53.

A portion of a shear cone is visible on column KK-3 in the same figure.

The south edge of the garage was supported by columns which were independent of building A-5. This was in contrast to the corbel support system

on the north edge.

The progress of formwork and stressing in the parking garage has not

been established at the time of this writing. These factors must be determined prior to analysis of the garage since they have a direct bearing on the behavior of the structure.

This unbonded post-tensi oned concrete construction such as used in the

parking garage, is a common form of construction. However, consideration of

its failure was beyond the scope of this report.

2.4 Crane No. 2

At the time of the collapse, crane No. 2 was reported idle with the jib

pointed towards the garage. The exact orientation is not known. A distant

view of the tower sections of the crane after the collapse may be seen in

figures 2.49 and 2.51 (also 2,5), The tower top and hoist machinery (see

figure 2,29) may be seen near column DD-2 in figure 2.55 (Note the partial

shear cones on column DD-2). Based on an examination of the photographs it

appears (refer to figure 2.1) that the crane tower lines up with the original

position of the crane base (in plan view) and passes almost over columns 32

and DD-3, ending up with the top as shown in figures 2.55 and 2.56. This

is a total distance of 125 to 130 ft. This distance compares to a total

tower height (base section, 5 standard sections, and 1 slipring section) of

about 125 ft. The implication here is that the tower fell while rotating

about its original plan position. However, careful inspection of figures

2.54, 2,55, and 2,5 shows the jib is lying alongside the tower sections.

The sequence of falling which would permit a jib initially pointing over the

garage (as reported by employees) to assume its final position has not been

determi ned

,

51
Figure 2.49 View of garage area looking northwest showing

crane no. 2 and column corbels on building A-4.
]

I

Figure 2.51 View looking west of area between the ramp and

building A-4.

Figure 2.52 View of ramp area looking south.

53

Figure 2,55 Position of fallen crane near column DD-2

looking south.

I Figure 2.56 Position of fallen crane near column DD-2

looking east. j

55

CHAPTER 3 CO NCRETE AND REINFORCEMENT IN BUILDING A-4

3.1 Concrete

3.1.1 Standard Cylinder Tests

The standard (ASTM C39) cylinder test results for compressive strengths

available at the time of this writing, are shown in figure 3.1 for the 17th

and 21st through 24th floors [18]. Both 7-day and 28-day results are shown

for section 3, the failure area. The results for section 3 indicate concrete

meeting the strength requirements called for by the structural engineer (3000

psi for the slabs and columns). The results of standard cylinder tests,

however, do not reflect the effect of field curing conditions or the strength

of the concrete in place. Field-cured cylinders were not made as called for

by OSHA regulations (ANSI A10.9, section 6.4.7) [2]. It should also be noted

that the test results for the 22nd floor, section 2, and the 23rd floor,

section 1, had strengths less than those normally obtained for other floors

at 7 days. Both of these sections were outside the failure zone.

3.1.2 Field Cores

In order to evaluate the strength of the concrete in the building the

portion of the slabs in section 3 between grid lines H and I (figure 2.1) were cored. Four-inch diameter cores were obtained from the structure in the

17th and 21st through 23rd floors. The results of compressive and split

cylinder tests on the cores are shown in figure 3.2 [17]. The 17th floor was sampled because crane no. 2 was originally thought to be supported on this

floor, although in section 2.2.3 it was concluded that the tower was based

on the 14th floor. The other floors were sampled because of their proximity

to where the failure started.

3.1.3 Estimates of Quality of Concrete as Delivered

The concrete quality as indicated by the standard 6-in diameter cylinders

was satisfactory for the 22nd and 23rd floors in section 3 (figure 3.1).

The field cores were used to check on the quality of concrete as delivered.

This comparison is made by converting the average values of 4-in diameter

core strength to 6-in standard cylinder values.

There has been conflicting information on whether strengths derived from cores are greater or lower than those from standard cylinders [3, 4, 8, and

16]. The strength of cores obtained from mass concrete (such as dams) frequently are greater than strengths derived from standard cylinders. However,

Bloem [4], using 6-in thick slabs with normal weight aggregate concrete, has

shown that 4-in diameter cores, such as obtained from building A-4, have lower

strengths than standard 6-in diameter cylinders. Earlier Campbell and Tobin

[8] obtained similar results using lightweight aggregate concrete and 4-in cores obtained from 12-in thick slabs. The data from Bloem is reproduced

in figure 3.3 to show the relationships obtained. Campbell and Tobin data

are similar but are not shown because of insufficient data at early ages.

Bloem tested cores from both well-cured and poorly-cured slabs. His

well-cured slab was sprayed with a curing compound as soon as the water

sheen had disappeared. Later the slab was covered with wet burlap and a plastic sheet. The burlap was kept wet and in place for 14 days. At that

time it was removed and the slab raised to permit air circulation. The

poorly-cured slab was left uncovered after pouring. Three days later the

57
Columns Cast

7 day cylinder strength

28 day cylinder strength

Slab Cast

7 day cylinder s trength

28 day cylinder strength

Columns Cast

7 day cylinder

28 day cylinder

1 trength

: trength

Slab Cast

7 day cylinder strength

28 day cylinder strength

ColumnB Cast

7 day cylinder strength

28 day cylinder strength

Slab Cast

7 day cylinder strength

28 day cylinder strength

Columns Cast

7 day cy Under strength

28 day cylinder strength

Slab Cast

7 day cylinder

28 day cylinder

itrength

itrength

2/28/73

2095 psi

2/26/73

2775 psi

2/23/73

1380 psi

2/14/73

2220 psi

2/ 7/73

3185 psi

4405 psi

2/ 5/73 2690 psi 3195 psi

25th Floor

3/ 1/73 2825 psi

2085 psi

2l2\m

2740 psi

2/15/73

1540 psi

2/13/73

2610 psi

2/ 6/73 3160 psi

35 20 psi

Slab Cast

7 day cylinder strength

28 day cylinder strength

4

2/28/73

2695 psi 3610 psi

2/26/73

2290 psi

3405 psi

2/23/73

2460 psi 3125 psi

2/20/73

2280 psi 3210 psi

2/14/73

2325 psi 3385 psi

2/ 7/73

2110 psi

3 025 psi

24th Floor

3/ 1/73 2325 psi

3235 psi

2/27/73

2575 psi 3750 psi

2/23/73 2460 psi 3125 psi

2/21/73

2115 psi 3205 psi

22nd Floor

2/15/73

2210 psi 3565 psi

2/13/73

2070 psi 3345 psi

21st Floor

1/10/73

3105 psi 4270 psi

1/11/73

2185 psi 3130 psi

Section 1 Section 2 Section 3 Section 4 Figure 2.1 Casting dates and laboratory oylinder strengths

Figure 3.2 Results of strength tests on 4-in

diameter cores.

Floor

No.

Number

of Cores Tested

Age at Time of Collapse,

Days

Age at Time of Test, DaysAverage

Strength at Time of Test,

ps i

Range of Strength

psi

Average

Density,

Ib/cu ft

4" Diameter Core Compression Tests

23 12 4 16 2295 1690-3360 119

22 13 10 25 1960 1390-2840 118

21 12 23 38 2470 1810-3150 117

17 12 51 66 27 80 2450-3420 124

4" Diameter Core Split Cylinder Tests

23 3 4 23 295 255-320

22 3 10 29 320 305-335

58
forms were stripped and the slab was raised from the floor. The environmental

conditions in the period between placement and coring of the slabs ranged in

temperature from about 60 to 90°F and in relative humidity from about 25 to

90 percent. The field curing conditions for each floor in building A-4 are not known.

Estimates were obtained for a range of standard 6-in diameter cylinders

by converting from the age at time of test to 28-day 6-in cylinder strength

using the curves of figure 3.3. The calculations rounded to the nearest 5 psi follow:

1 / T, n T 1 0 0 r . Wcl l"L.LiicU biau 0 7 Q n / n Q ft ^ 1 1 a u 1 u 1

1

0vy u u d d y s ) - ?c. ftO JT RJ P s 1 poorly-cured slab 2780/0 72 ( f racti on 0 66 days ) = 3860 psi

21st floor: wel 1 -cured si ab 2470/0 90 (fraction @ 38 days ) = 2745 psi

poorly-cured si ab 2470/0 72 (fraction @ 38 days ) = 3430 psi

22nd floor: wel 1 -cured si ab 1960/0 82 ( f racti on @ 25 days ) = 2390 psi

poorly-cured slab 1960/0 68 (fraction 25 days ) = 2880 psi

23rd floor: wel 1 -cured si ab 2295/0 73 ( f racti on (3 16 days ) = 3145 psi

Poorly-cured slab 2295/0 .62 (fraction @ 16 days ) = 3700 psi

Within the limitation of the scatter observed in the core test results (see

figure 3.2), the above data, when compared to the standard 6-in cylinders

(figure 3.1) indicate that the concrete quality as delivered was generally

acceptable although concrete strength of the 22nd floor slab appears to be

low. The foregoing is not a measure of the strength of in-situ concrete which

is discussed in the next section.

3.1.4 Estimates of Concrete Strengths on the Day of Collapse

The many factors influencing concrete strength have been discussed else

where [3, 14, 16, and 23] and are not repeated here. However, it should be

pointed out that for the same concrete:

(1) Strength will decrease with a deficiency in curing

moisture after the initial set of the concrete.

(2) Strength potential will increase with sustained low

temperatures (above freezing), although strength will

be less at the usual test periods (7 and 28 days).

(3) Strength development will accelerate with high

temperatures, although the eventual maximum strength

will be less than in (2).

(4) Calcium chloride will accelerate the rate of strength

development although it will not prevent freezing.

59
A set of curves have been

Klieger [14]. These curves, a concrete, indicate the effect

chloride on concrete compressi

curve without calcium chloride

ditions. Note the acceleratin

calcium chloride. For example

calcium chloride is equivalent

same concrete cured at 73°F.

calcium chloride was used for

drawn in figure 3.4 from data obtained by

Ithough obtained from normal weight aggregate

of curing temperatures and 2 percent calcium

ve strength development. The 73°F curing

represents strength gain under standard con

g effect on strength develop-ment of 2 percent

, a concrete cured at 55°F with 2 percent

in strength (during the first 28 days) to the

Although not used in all floors, 2 percent

the 17th and 21st through 24th floor slabs.

The cores are used as a measure of in-situ strength. The Klieger data

require the use'of 28-day strengths. The core strengths for the 22nd and

23rd floors are converted to 28-day strengths by using figure 3.3. The basic

core data are multiplied by the ratio of the 28-day fraction to the fraction

at the date tested. These calculations rounded to the nearest 10 psi follow:

22nd floor: well-cured slab 1 960x0.85/0.82 (fraction (? 25 days) = 2030 psi

poorly-cured slab 1960x0.69/0.68 (fraction @ 25 days) = 1990 psi

23rd floor: well-cured slab 2295x0.85/0.74 (fraction 0 16 days) = 2670 psi

poorly-cured slab 2295x0.69/0.62 (fraction (a 16 days) = 2550 psi

In order to convert these strengths into estimates at the time of

collapse the curing temperature should be considered. The average temperature

history has been shown in figure 3.5 (enlarged from figure 2.3). The

average air temperature was 42°F for the 22nd floor and 45° for the 23rd floor

(temperatures as recorded at National Airport). Estimates of strength using

the Klieger data for 2 percent calcium chloride in figure 3.4 (and interpolating between temperature curves) yields the following results rounded to the

nearest 10 psi:

22nd floor: well-cured slab

poorly-cured slab

23rd floor: well-cured slab

poorly-cured slab

2030x0.66 (fraction (a 10 days)

1990x0.66 (fraction @ 10 days)

2670x0.45 (fraction @ 4 days)

2550x0.45 (fraction @ 4 days)

1340 psi

1310 psi

1200 psi

1150 psi

In the prediction of concrete strength it is recognized that the average

in-situ strength of the floor slabs in question may be different because of

non-uniform curing temperatures, effects of coring and size of core, scatter

in test results in seemingly identical core specimens (see figure 3.2),

chilling factor of the wind and approximations introduced by application

of test results from independent sources of study. Consideration of these

factors, including quantitative assessment whenever possible, indicates

a range in the order of +_ 20 percent on the average concrete strength estimate.

The structural engineer's calculations for evaluation of shear strength

for the structure requires the use of a splitting ratio [6] for the lightweight

concrete. The structural engineer's design calculations (page 26A, floor

slab design for building A-4) provided data indicating that a splitting ratio

of 6.0 could be used. The splitting ratio is equal to the ratio of splitting

tensile strength to the square root of the compressive cylinder strength

(figure 3.2). The core data in figure 3.2 indicate a splitting factor of

6.5 to 7.0.

60
I

3 4 7 10 16 2 5 28 38 66 91 364

TIME, DAYS (log scole)

j Figure 3.3 Strength measures for concrete cores and cylinders . I

i

i

1 Figure 3.4 Development of concrete strength under ' various curing temperatures. |
3 . 2 Reinforcing Steel

Examination of the reinforcing steel in the area bounded by columns

66-68-83-86, indicates that the steel called for by the structural engineer

was included in the reinforcing steel shop drawings. The structural engineer

generally called for no. 4 bars in the slab. Slabs were not available for

examination to check if steel was actually placed in accordance with shop

drawi ngs

.

A total of 12 reinforcing bars were tested, 6 of these were removed

from floor slabs of the collapsed section of the building and 6 were tested

from unused reinforcement steel at the construction site. The results of

these tests are shown in figure 3.5 [17].

The structural engineer generally called for reinforcing steel with a 60,000 psi yield point and meeting the requirements of ASTM A-615. The

steel tested satisfies the requirements of that specification.

62
I

20 21 22 23 24 25 26 27 28 I \ 2 FEBRUARY | MARCH CALENDAR DATE Figure 3.5 Temperature history for section 3 on 22nd and

23rd floor slabs.

Figure 3.6 Results of reinforcing bar tests.

Reinforcing Bars Removed from Floor Slobs from Collopsed SecHon of Building Sample No. Size

Yield

Load, lbs. MaximumLoad, lbs. YieldStrength, psi

Tensile

Sliengfh, psi % Elongation

1 H 13,500 22,800 68,700 116,000 8.2

2 H 12,600 22,100 64, 100 112,500 8.9

3 H 14,000 21,600 71,300 1 1 0, 000 14.1

4 »5 22,000 34,000 71,700 110,800 9.4

5 »5 21,000 33,800 68,400 110,200 9.4

6 *7 38,400 62,000 63,800 103, 100 8.6

New Reinforcing Bars Obtained from' _ _ Jobsite Sample No. Size

Yield

Load, lbs. Maximum

Load, -lbs.

Yield

Strengtll, psi

Tensile

Sfreitgth, psi % Elongation

7 #4 13,500 21,750 68,700 110,700 13.3

8 #4 13,000 21,300 66,200 108.500 13.3

9 "5 21,400 34,800 69,700 113,400 13.9

10 *5 21,600 35,600 70,400 116,000 13.3

11 "7 39,500 64,200 65,700 109,300 11.2

12 *7 39,400 65,000 65,500 108,900 10.9

63

1

CHAPTER 4 STRUCTURAL INVESTIGATION OF FAILURE

CONDITIONS IN BUILDING A-4

4. 1 Introduction

A structural investigation of the 22nd and 23rd floors of the Skyline

Plaza apartment building A-4 in the region where the collapse occurred was conducted. The investigation consisted of:

(1) determination of internal forces in slab,

beam, and column elements of the structural

assembly in accordance with the principles

of elastic analysis;

(2) evaluation of the ability of structural

elements to resist previously determined

f orce s .

The three-dimensional elastic analysis was performed using the finite

element analysis program known by the acronym SAP [22A] and the computer

facilities at the National Bureau of Standards. The simulated model of the

structural assembly shown in figure 4.2 consists of three-dimensional beam

elements simulating beams and columns of the system and quadrilateral plate

bending elements simulating the floor slab. The beam element properties are discussed in the SAP user manual while the plate element properties are described in reference [9].

The evaluation of the capacity of structural elements was based on the

provisions of the ACI 318-71 Code [7] and on procedures from published analytical and experimental research [12, 13, 13A, 15].

4 . 2 Controlling Regulations and Basis of Design

The basic national standard governing the design of reinforced concrete

buildings is the Building Code Requirements for Reinforced Concrete by the

American Concrete Institute, ACI 318 [6,7]. New editions of the ACI Code

are issued periodically to provide for the advancement in the state of the

art resulting from research and professional experience. The latest edition

of ACI 318 was issued in 1971 [7] and the one prior to that, in 1963 [6].

In building design the applicable building code is the latest edition

available at the time of the design. In the case of the Skyline Plaza

apartment building, the design should comply with the requirements of the

1956 edition of Building Code Ordinance (amended in 1971) of Fairfax County,

Virginia [5], which incorporates the design provisions of ACI 318-63.

It is noted that for certain areas of the floor the spacing of columns

is such that in order to comply with section 2101(e)2 of ACI 318-63, the

slab thickness would have to be greater than 8 in. Section 3 of the floor

slab, as identified in figure 2.1 contains such areas.

4 . 3 Analytical Procedure

Whereas the design of the structure was governed by ACI 318-63, the

analysis that follows is based upon ACI 318-71 which represents the present

65
state of the art. The calculations of shearing stresses were made in accord

ance with the procedure described in section 11.13.2 of the Commentary to the

ACI 318-71. The latter is essentially based on the results of experimental

studies by Hanson and Hanson [13] of slabs supported by square columns. In addition, consideration was given to the effect of column rectangul ari ty on shear capacity as reported by Hawkins, Fallsen and Hinojosa [13A]. For

distribution of loads between shored slabs consideration was given to findings

by Grundy and Kabaila [12], Nielsen [15], and observations of field measurements of high-rise flat slab buildings related by Agarwal and Gardner in an unpublished report.

Three finite element analyses were made for the slab assembly in the

region of collapse. The approximate area of collapse is shown in figure 4.1.

All three analyses were based on the grid model shown in figure 4.2.

The accuracy of a finite element analysis is generally improved as the

number of elements is increased and as the elements become more square. In constructing the grid, a finer mesh was used in the region west of the crane tower opening, where failure was assumed to have been initiated. The

boundaries of the rectangular grid are approximately defined by the column

line 18-73-72-79-28 on the east and column line 11-63-64-90-91-35 on the west.

This area encompasses the failure region and extends about one column line

beyond its boundaries on each side.

The slab element thickness was 8 inches as called for in the contract

drawings. The supporting column elements were rigidly attached to the

slab elements at the top and were fixed at their base. The columns above

the floor being analyzed were also rigidly attached to this slab at their

base with the tops hinged. Columns designated by Gand 32 in the contract

drawings were each simulated by a rigid frame consisting of two columns and

astiffbeam.

The grid layout in figure 4.2 is different from the slab in the contract

drawings in certain aspects. The balcony slabs were assumed extending to the

centerlines of the end columns. Likewise the boundary of the slab at north

and south fascias between balconies was assumed extending to the centerlines

of the flanking columns, eliminating, in effect, the stepped portion of the

slab between a balcony and an adjacent fascia. Wherever possible, rectangular

or square plate elements were used to construct the grid. In order to make

column centroids coincident with nodal points of the grid a few trapezoidal

plate elements were also introduced. By using a finer mesh one could develop

a model which would more precisely represent the actual assembly. However,

such a refinement is not justified because it would not significantly affect

the results [7A].

Each of the three finite element analyses had a particular purpose:

Case 1 : Some of the employee statements indicated that the

22nd story forms were entirely removed at the time

of collapse (see section 2 .-2 . 2 ) . This analysis of

the 23rd floor system is based on the premise that

the 23rd floor slab was not shored. The loading

consisted of the weight of the newly poured slab of

the 24th floor (80 pounds per square foot), the weight

of the formwork on the 23rd floor (5 pounds per square

foot) and the weight of the 23rd floor (80 pounds per

square foot). Thus, a total uniform load of 165 pounds

per square foot was placed on the portion of the slab

. directly below the poured section of the 24th floor

(section 3) and a load of 85 pounds per square foot

66
o

CO

COOO

«^

CM

03CO

<M

O

-Pc o « a

CO

Snoo E o HO «

03

is

67
was placed elsewhere. The modulus of elasticity

used for the concrete slab of the 23rd floor was calculated in accordance with ACI 318-71 using the

estimated average concrete strength of 1200 psi

(see section 3.'1.4), and an approximate density of

120 pounds per cubic foot (see table 3.2). A similar

calculation was made for the elastic moduli of the

columns. For stiffness calculations, the moment of

inertia was assumed to be that of the gross section

of the concrete.

Case 2 : To check the influence of estimated concrete strength

on the analytical results. Case 2 assumed theconcrete

to have attained its full 28-day design strength of

3000 psi. The 23rd-floor system was analyzed using

elastic moduli calculated as in Case 1 except that

design concrete strength was used.

Case 3 : Some of the employee statements indicated that removal

of shoring under the 23rd-floor slab in the central

corridor area of the building was in progress on the

day of the collapse (see section 2.2.2). A loading

simulating the condition of partial removal of shoring

was used. This Case is similar to Case 1 except the

loads on the 23rd floor slab were reduced outside the

region bounded by grid lines 7-320, 320-330, 330-16,

and 16-7 as indicated in figure 4.2. To obtain these

reduced loads, the loads from case 1 were multiplied

by a factor which was based on the assumption that

22nd and 23rd floor slabs shared the loads on the 23rd

floor in proportion to their respective elastic moduli.

The modulus of elasticity of the 23rd floor slab was the same as was used in Case 1. The modulus of elasticity

of the 22nd floor slab was calculated on the same basis

as described in Case 1 using the average estimated concrete

strength at the time of collapse of 1340 psi for the 22nd

floor slab (see section 3.1.4).

4 . 4 Discussion of Results

Figures 4.3 through 4.5 show the analytical results for cases 1 through

3. Moments M (rotation about x-axis) and M (rotation about y-axis) are designated by^hor i zontal and vertical 1 ines , ^respectively. They are shown

plotted in pairs (and approximately to scale) at the centroids of slab elements. The numerical entries are moment magnitudes at the centroids and are expressed in units of foot^kips per linear foot of slab. Moments producing

compression in the top of the slab are positive.

Figures 4.6 through 4.9 show moment diagrams developed for column strips

taken from the grid of figure 4.2. These strips were investigated for their

adequacy to resist moments obtained by analysis. Negative moments were investigated at the face of the columns according to ACI 318-71 .

68

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9#4T

133 149 168 188 207 222 239 257 277 294 310

SECTION MOMENT CAPACITY

CASE

(CASE 2 SIMILAR)

CRITICAL SECTION

CASE 3, NOT CRITICAL Figure 4.6 Moments in column strip 32-32-31-30

.

Figure 4.7 Moments in column strip 15-83-31.

243(15)

12x24

246 247 248 249 250 251 252 253 254 255 256 257

167

, I6»24

.258(31) 261 12 83

170 i '^, -,14152

1

I . 171

1

I 125'

SECTION IVIOMENT CAPACITY

C

'2 ^ -4 « -6

-2,6 -3,7

4#4 a l#5

\ CASE I

' (CASE 2 SIMILAR)

73
A

I6i24

211 212 213 214 215 216 217 218 219 220 221 \ 130 131 132 133 134 135 136 137 138 139 140 141 142 143

195 196

197 681 i,.4, 200 201 202 203 204 205 206 207 208

191(141/

/ 1 16 117 118 119 120 121 122 123 124 125 126 127 128 129

/|76 177 178 179 180 181 182 183 184 185 186 187 188 189

4,83'

209|

5.84' 190 i

WEST STRIP

• EAST STRIP • SECTION MOMENT CAPACITY

Figure 4.8 Moments in aolumn strip 191-197.

Figure 4.9 Moments in column strip 13-84-32.

J 177 178 179 180 181 182 183 184 185 186 187 188 189

74
The slab was also investigated at column locations for its adequacy in shear. Shearing stresses were determined following the procedure described

in section 11.13.2 of the commentary to the ACI 318-71 code. The applicable equations are given below for convenience. Figure 4.10 shows the critical section for shear around the periphery of a rectangular column and other parameters used in the following equations:

^max A J ^ 2 "j— ^T'' (4.1)

y X

J

_ d(b+d)^ ^ d^(b+d) d(h+d)(b+d)^

X 6 6 2

J

= d(h+d)^ ^ d ^(h+d) , d(b+d)(h+d)^

y 6 6 2

a. . = 1 -

1

1 +

1

2

/ b+d

3 Vh + d

y H

A = 2d (b+h+2d)

where

v = maximum shear stress in the slab

max

M = bending moment about x-axis transmitted to column

X

^ My = bending moment about y-axis transmitted to column

b,h = sectional dimensions of rectangular column

d = distance from centroid of tensile reinforcement

to bottom face of slab

a ,a = moment reduction factors defined above

X y J^,Jy= properties of slab section as defined above

V = vertical load transmitted to column

A = area of critical slab section peripheral to column

as defined above

According to ACI 318-71, the limiting value for shear stress is 4 'for normal weight concrete (section 11.13.2) and 85 percent of that valu§ or 3.4 -/P" for "sand-lightweight" concrete (11.3.2) such as used in the floor

slabs. ^ Critical shear in the slab will occur when the maximum shear stress

determined by equation (4.1) exceeds 3 . 4 Vf^ or,

V > 3.4Vn" (4.2) max * c ^

'

75
Alternately, expression (4.2) may be solved to give the minimum concrete

strength required to prevent shear failure,

Therefore, the shear capacity of the slab is considered to be exceeded when

the required compressive strength determined in accordance with equation

(4.3) is greater than the estimated strength at the time of collapse.

In addition to flexure or shear type failure in the slab, the collapse

could have been attributed to other causes such as excessive creep in concrete

or failure of one or more of the supporting columns. A limited investigation

indicated that the columns had sufficient capacity to resist the applied loads.

The possibility of creep being a significant contributory factor in the

collapse was ruled out because of the short duration of peak loads on the

23rd and 22nd floor slabs prior to the collapse (about 4 hours) and the

presence of reinforcement in the compressive regions of the slabs.

The scope of the following investigation is based on the hypothesis that

failure conditions were governed either by flexure or by shear.

A . Investigation of Flexural Failure

A comparison of results shown in figures 4.3 and 4.4 reveals the

internal moments in the slab from the analyses of Cases 1 and 2 to be

virtually identical. This indicates that the accuracy of predicted

concrete strength will not have a significant influence on the results

of flexural investigations.

Altogether four column strips were investigated. Of these, the

strip shown in figure 4.6 was found to be the one most critical in

flexure. Case 1 negative moment at the critical section near column

31 (node 258) was about 9.2 ft-kips/ft. The moment capacity of the

slab at this section, determined on the basis of rein-f orcement details

in the contract drawings, was found to be about 8.4 ft-kips/ft, indicating that the flexural capacity of the section was exceeded by about

10 percent. In this case, as in all other cases investigated, the

flexural capacity of the section was found to be governed by yielding

of the tensile reinforcement rather than by crushing of concrete in

compression. Thus it is estimated that yielding occurred in the slab

at column 31 and yielding propagated toward column 83 (see figure 4.1).

However, this local yielding is not sufficient to cause a collapse

mechanism of the 23rd story floor slab. In fact, a flat slab construction of the type used in Skyline Plaza building is generally recognized

to have considerably greater moment capacity than the capacity at which

the first local yielding occurs because of its inherent ability to

redistribute peak moments to neighboring regions through the mechanism

of yielding.

The next most severe condition occurred in the column strip shown

in figure 4.9. Maximum negative moments in elements adjacent to column

84 (node 165) were in the order of 7 to 8 ft-kip/ft at the critical

section. This compares with a 10 ft-kip/ft ultimate moment capacity

in the slab. The difference is not large enough to preclude yielding

or even some yield line propagation from column 84 in the east-west

direction; however, if such yielding were to occur the situation would

be very similar to the localized condition around column 31. Again, it

is concluded that this was not a significant contributory factor to the

collapse of the 23rd floor.

76
To further check the ultimate flexural capacity of the floor slab

a selected critical area was investigated using yield line analysis.

The yield line pattern assumed for the balcony panel bounded by columns

32, 84, 83 and 31 is shown in figure 4.11. Top and bottom reinforcement

taken from the contract documents is indicated in the same figure. This

particular analysis gave an ultimate load capacity of about 300 pounds

per square foot indicating 80 percent excess capacity over construction

loads. Other yield line analyses performed at selected interior panels

gave consistently similar results. It should be noted that unless the

assumed yield line pattern happens to be the same as would actually

develop in the slab, the approach used will overestimate the slab

capacity. However, the margin of approximation involved would be much

less than the excess capacity indicated by these analyses.

The possibility that the collapse might be attributed to flexural

failure in the 22nd story slab was also considered. The most critical

situation would occur when the 23rd story slab is fully shored and no reshores are installed under the 22nd floor slab (some employee statements

indicate that the 23rd floor slab could have been fully shored, as discussed in section 2.2.2). This condition will be designated hereafter

as Case 4. In this case, the loads acting upon the 22nd floor would

consist of its own weight, the weight of the 22nd story shoring and the

loads transmitted from the 23rd floor. Assuming the loads on the 23rd

floor to be distributed to both slabs in proportion to their respective

elastic moduli, and shoring to be perfectly rigid [12], the maximum

total load that could act on the 22nd floor would be about 4 percent

greater than those assumed acting on the unshored 23rd floor (Case 1).

Since both slabs are of identical design, the internal forces (shears,

moments and torques) in the 22nd floor slab would likewise be about

4 percent greater. However, the flexibility of timber shores would

modify the load distribution between slabs significantly [15], with the

net effect of reducing the loads transmitted to the 22nd floor (viz.:

a 100 percent flexible shore transmits no loads). Further reduction in

transmitted loads would occur as a result of poor shoring conditions

which appears to be a reasonable assumption to make (see section 2.2.2).

Taking all these factors into account, it is concluded that flexure in

the 22nd floor slab did not constitute the initial mode of failure.

This conclusion is further corroborated by observations made elsewhere

in this report (see section 2.2.2 and figure 2.28). It should be noted

that since the 23rd floor slab, when fully shored, is less critical in

flexure than either in Case 1 (no shoring) or in Case 3 (partial shoring),

it requires no consideration with respect to Case 4. Upon examination of all the probable conditions prior to the

collapse, it is concluded that the initial mode of failure of either

the 22nd or the 23rd floor slab was not flexural.

B . Investigation of Shear Failure

From the analytical results, axial loads and moments transmitted to column elements from the floor system were assessed in relation to

peripheral column dimensions in order to establish which of the columns

would be expected to create a severe shear condition in the slab.

Figure 4.12 shows the four most critical regions identified at columns

67, 68, 83 and 84.

77
Y

d/2

b-i-d 4- . M, i—^

d/2

htd

Figure 4. 10 Typical reotangular aolumn showing critical

section for shear.

Figure 4.11 Yieldline pattern in balcony panel.

7#5T i I ll#^T

I

78
18 5

r

30 5

24

Mx= 21.7

1 My=433'

16

22.5

Column 67

(a)

36 5" r

54 5

Column 68

(b)

22.5

30 5

Column 83

(d)

Figure 4.12 Cvitiaal seation for shear in the slab

around columns 67^68,85^84.

79
Figure 4,13 tabulates the results of the analysis of shear stresses

in the 23rd-story floor slab for cases 1 through 3, based on the critical

sections shown in figure 4.12. Maximum shear stresses calculated in

accordance with eq . (4.1) are given in column 3. Column 4 gives the

required compressive strength to resist this shear as determined from

equation (4.3) and column 5 gives the probable range of in-situ concrete

strength at the time of collapse.

According to recent studies [13A] the maximum shear capacity of a slab is significantly reduced when the supporting column has a narrow rectangular cross section as compared to a square column. In the case

of column 68, the reduction in strength would be in the order of 20

percent. The bracketed values in column 4 of the table reflect the

effect of column rectangul ari ty

.

Examination of the tabulated results reveals that the shear capacity

of the 23rd floor slab is exceeded at columns 67, 68, 83, and 84 under

unshored and partially shored conditions (Cases 1 and 3, respectively).

It can be seen by comparing Case 1 values in column 4 with the probable

range of compressive strength in column 5 that each of the four regions

of the slab at the columns is critical in shear. For Case 3, conditions

are critical at columns 84 and 68 and marginal at columns 83 and 67.

Shear stresses at these columns are comparable to stresses in Case 1,

indicating that partial removal of shoring creates conditions nearly as critical as would develop in the unshored slab. Note that none of the

strength requirements exceed the 3000 psi design strength of the concrete

considered in Case 2. The loss of support as a result of shear failure

at any one of these columns would cause a progressive propagation of

failure at neighboring columns and eventual rupture of the slab.

The cases discussed above represent conditions under which the 23rd

story slab is the one most critically stressed in shear. The possibility

of shear failure having first been initiated in the 22nd story slab was also examined. The most critical condition in this slab would occur under Case 4 loading (23rd floor fully shored and 22nd floor not reshored).

As previously noted, with infinitely rigid shoring, the load acting

on the 22nd floor in Case 4 would be about 4 percent greater than the

Case 3 load acting on the 23rd floor. However, the actual load distribution between the two floors would be significantly modified by the

flexibility of the timber shoring [15] with the net effect of reducing

the load transmitted to the 22nd floor. The load transmitted would be' decreased further by the poor condition of shoring such as described

in section 2.2.2.

The effect of these two factors on the load distribution cannot be

quantified because of insufficient information. However, the load

distribution which would simultaneously mobilize both floors towards

their respective shear capacities can be examined; a condition under

which the combined resistance of both floors is a maximum. In this

case, the total weight of three floors plus shoring in two stories would

be distributed between the 22nd and 23rd floors in proportion to the

square roots of their respective compressive strengths. The results

indicated required compressive strengths of 1030 psi and 1115 psi for

the 23rd and 22nd floors, respectively. These compare with 1200 psi and

1340 psi respective average values estimated for these floors. Consequently, failure under Case 4 loading cannot be precluded. However, the

80
Column No. Case

V

max

psi

c required

psi

£'c es t imated

psi

1 147 1870 960-1440

67 2 147 1870 3000

(16 X 24) 3 128 1417 960-1440

1 113 1104 (1724) 960-1440

68 2 113 1104 (1724) 3000

(48 X 12) 3 116 1164(1816) 960-1440

1 148 1894 960-1440

83 2 148 1894 3000

(24 X 16) 3 130 1462 960-1440

1 149 1920 960-1440

84 2 149 1920 3000

(30 X 12) 3 145 1819 960-1440

-^The values in brackets are based on v^^^ > 0.8

^ IliclA (3.4VTJ) or = 0.135 v^^^^ to account for the

effect of column rectangularity [13A]

.

Figure 4.13 Summary of shear stresses in

the 23rd story floor slab.

81
following evidence is taken as an indication that the collapse did not

initiate in the 22nd floor slab: (a) employee statements did not indicate

sagging of the 22nd floor slab at the time of collapse and (b) one employee

statement indicating the loss of the 21st story reshores prior to the incident

was interpreted as a decrease in the 22nd floor deflection (section 2.2.2

and f i gure 2 . 28 )

.

The contract documents (structural drawings) specify the following:

"Slab being poured to be shored for two floors and backpropped at center of

span each way and at center of bay on next floor down." Uncertainty about

the effectiveness of backpropping or its presence prior to the collapse

(section 2.2.2 and figure 2.27) makes it virtually impossible to make a quantitative assessment of loading distribution between 23rd through 21st

floor slabs for this shoring configuration.

Upon examination of all the probable conditions prior to the collapse,

it is concluded that the initial shear mode of failure of the 23rd story

floor slab resulting from partial or complete removal of shoring prior to the incident was a major contributing factor to the collapse.

4 . 5 Probable Mode of Collapse in Building A-4

The most likely mode of collapse has been determined to be a shear

failure around columns 57, 58, 83, or 84. The premature removal of forms

supporting the 23rd story slab when the concrete of that slab had a relatively

low strength produced shear stresses that were in excess of the concrete

capacity at the time of the incident.

The three-dimensional finite element analyses have shown the slab to

be overstressed in flexure in only a few local regions. The capacity of

flat slabs to redistribute moments is well known and thus local flexural

yielding should not have led to failure. Approximate ultimate flexural

capacities were computed by the yield line analysis method. These ultimate

flexural loads indicated that, even with the forms removed, the 23rd floor

slab should not fail in flexure, thus confirming the interpretation of the

results of the elastic analyses.

Hawkins test results [13A] indicate that shear cracking can cause concentrated slab rotations near the column that are quite large. Based

on this observation it is felt quite possible that large deflections could

have occurred, even with a shear type of failure that is estimated to have

occurred. Most of the eyewitness reports indicated deflection in the 23rd

and 24th story slabs (varying from 6 in to 2 ft) which increased over a 15 or 20 minute time period before failure. An increasing deflection of this

type is usually associated with a flexural failure; however, this type of

deflection could also be associated with a shear failure to the layman

observer.

The collapse is believed to have started with shear around columns 67,

68, 83, or 84. The loss of support from any one of these columns would then

lead to overstress i ng of the remaining columns.

The accumulation and impact of debris from the 23rd and 24th floor slabs

would have overloaded the 22nd floor slab and induced progressive collapse

of successive floors to the ground.

82
There is no indication that the crane was a contributing factor to the

beginning of collapse in building A-4. No witness statements indicated that

the crane moved prior to or during the initial sagging of the 23rd and 24th

floors. The crane supports on the 14th and 16th floors are far enough away

from the initiation of failure to preclude the crane as a cause. However,

the crane probably became a driving force in the collapse once its support

was lost.

83

CHAPTER 5 SUMMARY OF FINDINGS

The findings given in this section are based on site investigations,

OSHA case records, structural and architectural drawings, shop drawings, and

structural calculations. The applicable Federal regulations are the Safety

and Health Regulations for Construction [10] which incorporates the American

National Standard A 10.9, Safety Requirements for Concrete Construction and ' Masonry Work [2]

.

5 . 1 Mode of Failure

On the basis of evidence as well as analysis, it appears that the coll- i apse was initiated at the 23rd floor level.

An analysis of the 23rd-floor slab indicates that its most likely mode ' of failure was in shear around one or more columns in section 3 of the floor

slab. The strength of the 23rd-floor slab on the day of collapse has been

found to be of a magnitude that complete or partial removal of shoring underneath the slab would have produced a shear failure in the slab. The weight

of debris resulted in failure in the slabs below and carried through the i height of the building.

5 . 2 Non-Compliance with OSHA Regulations

I Non-compliance with OSHA regulations was found in a number of instances.

These are listed below along.with a discussion of each item.

(a) Shoring in section 3 of the 22nd story

Examination of physical evidence and employees' statements indicate that i the 22nd story forms were being removed on the day of the incident. OSHA ' regulations (ANSI-A10.9, section 6.4.7) require adherence to engineer's

specifications and local building codes in determining length of time for ' forms to remain in place. The engineer's requirements were expressed in the

form of a note on the structural drawings (section 2.2.2). This note required

the "slab being poured to be shored for two floors and backpropped at center i of span each way and at center of bay on next floor down." The architect's , specifications [19] reauired "In all cases, two floors shall be fully shored." ! The removal of the 23rci story forms left only one story of formwork in place j under the recently poured 24th floor. |

i (b) Premature removal of 22nd-story forms i

The length of time forms were required to be left in place was not ! explicitly stated by the engineer, architect or local code. In such instances j OSHA regulations (ANSI-A10.9, section 6.4.8) provide minimum curing times. | When the design live load is less than the dead load, 4 days are required

for spans less than 10 ft, 7 days for 10 ft to 20 ft spans, and 10 days for

spans exceeding 20 ft. The time periods are for cumulative numbers of days ' in which the air temperature surrounding the concrete exceeds 50°F. The

4-day old 23rd floor slab had spans exceeding 10 ft. The forms removed on the 22nd story were in an area with spans exceeding 20 ft and therefore,

should have been in place for 10 days of temperatures exceeding 50°F. | (c) Field-cured concrete specimens

OSHA regulations (ANSI-A10.9, section 6.4.7) require the use of fieldcured concrete specimens in order to insure that concrete has obtained suffi- i cient strength to safely support the load prior to removal of forms. No

evidence has been found which indicates that field-cured specimens were prepared or used.

85

i
(d) Lateral bracing

OSHA regulations (ANSI-A10.9, section 6.3.2 and 8.1.5) require the

design of braces and shores to resist all foreseeable lateral loads.

Minimum values of 100 pounds per foot of floor edge or 2 percent of the

total dead load of the floor, whichever is greater, is required. No

evidence has been found which indicates that lateral load was considered in

the design of forms. The lateral bracing provided (about 2 nominal 3x4's

per 16 ft) would not provide this resistance.

(e) Shoring out of plumb

OSHA regulations (ANSI-A10.9, section 8.1.24) allow a maximum deviation

of 1/8 in per 3 ft out of plumb. Deviations of shoring exceeding these

limits were found on the 23rd (figure 2.20) and 24th (figure 2.15) stories.

(f) Damaged shoring

OSHA regulations (ANSI-A10.9, section 8.1.25) require removal of damaged

or weakened shoring. On-site inspection after the incident indicates this

was not done on the 23rd (f igure 2.22) and 24th (figure 2.18) stories.

(g) Inspection

OSHA regulations (29 CFR, 1926.700 (e) (1) (iv) ) require inspection

immediately before, during, and after placing concrete. This either was not

done or deficiencies in (e) and (f) above were not corrected.

(h) Crane Installation

OSHA regulations (29 CFR, section 1518.550, 6, 5) require the operation

of cranes to be as prescribed by the manufacturer. The following deviations

were found:

(1) Crane no. 1 (in section 1, away from the collapse

area) - distance between top and bottom supports

was less than the required 18 ft 4 in (figure 2.30)

(18 ft was used ) .

(2) Crane no. 2 (in section 4, in the collapse area)-

distance between top and bottom supports was less

than the required 21 ft 4 in (18 ft was used).

(3) Crane no. 2 (in section 4, in the collapse area) - The number of standard tower sections used was one more than the four recommended by the manufacturer

(figure 2.41 and section 2.2.3).

(4) Crane no. 2 (in section 4, in the collapse area) - The maximum tower height exceeded (by approximately

one standard tower section) the 81 ft-0 in recommended

(figure 2.31 and section 2.2.3).

86
CHAPTER 6 REFERENCES

1. ACI-ASCE Committee 326, "Shear and Diagonal Tension," Journal of the

American Concrete Institute , Vol. 59, January, February, and March

2. American National Standard Institute, "Safety Requi rementsf or Concrete

Construction and Masonry Work," ANSI A10.9- 1970, American National

Standard Institute, Inc., New York, New York, 1970.

3. Bloem, D.L. "Concrete Strength Measurement--Cores versus Cylinders,"

Proceedings of the American Society for Testing Materials , Vol. 65,

1 965.

4. Bloem, D.L. "Concrete Strength in Structures," Journal of the American

Concrete Institute , Vol. 65, March 1968.

5. Building Code Ordinance, County of Fairfax, Virginia, Adopted May 2, 1956

with Amendments to August 4, 1971.

6. Building Code Requirements for Reinforced Concrete (ACI 318-63) , American

Concrete Institute, Detroit, Michigan, June, 1963.

7. Building Code Requirements for Reinforced Concrete (ACI 318-71) , American

Concrete Institute, Detroit, Michigan, 1971.

7 A . The Building Collapse at 2000 Commonwealth Avenue, Boston, Massachusetts , Report of the Mayor ' s Investigating Commission, City of Boston,

Massachusettes ,1971

8. Campbell, R.H. and Tobin, R.E. "Core and Cylinder Strengths of Natural

and Lightweight Concrete," Journal of the American Concrete Institute , Vol . 64, April 1 967.

9. Clough, R.W. and Felippa, C.A. "A Refined Quadrilateral Element for

Analysis of Plate Bending," Proceedings, Second Conference on Matrix

Methods in Structural Mechanics , Wright-Patterson Air Force Base, Ohio,

1 968.

10. Department of Labor, Bureau of Labor Standards, "Safety and Health

Regulations for Construction," (29 CFR Part 1926), Washington, D.C.

11. Formwork for Concrete SP-4 , American Concrete Institute, Detroit,

Michigan, 1963.

12. Grundy, P. and Kabaila, A. "Construction Loads on Slabs With Shored

Formwork in Multistory Buildings," Journal of the American Concrete

Insti tute , Vol. 60, December 1 963.

13. Hanson, N.W., and Hanson, J.M. "Shear and Moment Transfer between

Concrete Slabs and Columns," Journal PCA Research and Development

Laboratory, Vol. 10, No. 1, January 1968.

13A. Hawkins, N.M., Fallsen, H.B. and Hinojosa, R.C. "Influence of Column

Rectangul ari ty on the Behavior of Flat Plate Structures," Cracking,

Deflection, and Ultimate Load of Concrete Slab Systems , SP-30, American

Concrete Institute, 1971.

87
14. Klieger, P. "Effect of Mixing and Curing Temperature on Concrete

Strength," Journal of the American Concrete Institute , Vol. 54, June

1 958.

15. Nielsen, Knud, E.C., "Loads on Reinforced Concrete Floor Slabs and Their

Deformations during Construction," Proceeding no. 15 Swedish Cement and

Concrete Research Institute, Royal Institute of Technology, Stockholm,

Sweden, 1952.

15. Price, W. H. "Factors Influencing Concrete Strength," Journal of the

American Concrete Institute , Vol. 47, 1951.

17. "Results of Core Testing Project at Skyline Center Building No. 4,"

Value Engineering Laboratory Report No. 45-RD-73, Alexandria, Virginia,

March 21 , 1 973.

18. Skyline Plaza Collapse - Case Records , Occupational Safety and Health

Administration, Region 3, Richmond, Virginia, 1973.

1 9 . Specifications for Skyline Apartments A-4, Fairfax County, Virginia , Wei he , Black, and Jeffries, Architect, Washington, D . C . , September 5 ,

1 972.

20 . Specifications for Structural Concrete for Bui 1 di ngs , ACI 301 -72 (with

selected ACI and ASTM references) SP-15C72) , American Concrete Institute

,

Detroit, Michigan, 1972.

21. Tower Crane Manual PC-14Q0 , American Pecco Corporation, Millwood, New

York.

22. Tower Crane Manual PC-1600 , American Pecco Corporation, Millwood, New

Yor k

.

22A. Wilson, E.L. SAP, A General Structural Analysis Program," Report to

Walla Walla District U.S. Engineers Office, September 1970 (not

publ i shed ) .

23. Winter Concreting, Theory and Practice , Proceedings of the RILEM

Symposium, Copenhagen, 1956.

88
|^trSS--TT4A (REV. 7-7 3)

U.S. DEPT. OF COMM. BIBLIOGRAPHIC DATA

SHEET

1. PUBLICATION OR REPORT NO. BSS 94

2. Gov't Accession

No.

3. Recipient's Acces.sion No.

4. TITLE AND SUBTITLE

Tr»TTcici"no'a1"Tr»r> n"F "hViP Slfv1"fnp Pla^P nnllpTiQP "in Fairfax County, Virginia

5. Publication Date February 1977

6. Performing Organization Code

^''^^gar%V Leyendecker and S. George Fattal 8. Performing Organ. Report No.

9. PERFORMING ORGANIZATION NAME AND ADDRESS

NATIONAL BUREAU OF STANDARDS

DEPARTMENT OF COMMERCE

WASHINGTON, D.C. 20234

10. Pro iect /Task/Work Unit No. 460 5503

11. Contract/Grant No.

12. Sponsoring Organization Name and Complete Address (Street, City, State, ZIP)

Occupational Safety and Health Administration

U.S. Department of Labor

Washington. D.C. 20210

13. Type of Report & Period Covered

Final/June 1973

14. Sponsoring Agency Code

15. SUPPLEMENTARY NOTES

Library of Congress Catalog Card Number: 77-600002

16. ABSTRACT (A 200-word or less factual summary of most si^ificant mformation. If document includes a significant

bibliography or literature survey, mention it here.) The collapse of the Skyline Plaza apartment building A-4 has been studied by using

information contained in case records of the Occupational Safety and Health

Administration (OSHA) , U.S. Department of Labor and obtained from on-site inspections

by investigators from the National Bureau of Standards

Non-compliance with OSHA construction standards has been identified with regard to

formwork, field-cured concrete specimens and crane installation. Specifically, the

construction procedures did not comply with standards for the removal of supporting

forms. It is concluded that premature removal of forms was a contributing factor to

the collapse in building A-4.

An analysis of the 23rd-floor slab indicates that its most likely mode of failure

was in shear around one or more columns in section 3 of the floor slab. The strength

of the 23rd-floor slab on the day of collapse has been estimated to be at a level that

removal of shoring could have produced shear failure in the slab

.

17. KEY WORDS (six to twelve entries; alphabetical order; capitalize only the first letter of the first key word unless a proper name; separated by semicolons) Apartment building; collapse; concrete; concrete strength;

construction; flexure; progressive collapse; shear; strength

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