A 11 ID 2 ATD134

MBS

PUBLICATIONS

NBSIR 88-3788

Inelastic Behavior of Full-Scale

Bridge Columns Subjected to

Cyclic Loading

U.S. DEPARTMENT OF COMMERCE

National Bureau of Standards

National Engineering Laboratory

Center for Building Technology

qq- —, MD 20899

100

U56 ! 88-3788

1988

C . 2

TMENTOF COMMERCE

IEAU OF STANDARDS

75 Year* Stimulating America's Progress

1913-1988

NBSIR 88-3788

INELASTIC BEHAVIOR OF FULL-SCALE

BRIDGE COLUMNS SUBJECTED TO

CYCLIC LOADING

William C. Stone

Geraldine S. Cheok

U.S. DEPARTMENT OF COMMERCE

National Bureau of Standards

National Engineering Laboratory

Center for Building Technology

Gaithersburg, MD 20899

May 1988

U.S. DEPARTMENT OF COMMERCE, C. William Verity, Secretary

NATIONAL BUREAU OF STANDARDS, Ernest Ambler, Director

0

°*eA\j of

ABSTRACT

Circular, spirally reinforced concrete bridge columns were subjected to

cyclic loading (representing seismic loads) in the laboratory. The test

articles were prototype columns designed in accordance with recent [1983]

California Department of Transportation (CALTRANS) specifications.

Two full-scale columns, each measuring 5 feet (1.52m) in diameter with aspect

ratios (height/width) of 3 and 6, were subjected to slow reversed cyclic

lateral load with constant axial load to simulate the gravity weight of the

bridge superstructure. Details are presented concerning the design of the

special test apparatus required to conduct the tests as well as

recommendations for future improvements in test procedures

.

Results from the tests are presented in the form of energy absorption graphs

and bar charts, load-displacement hysteresis curves, longitudinal and

confining steel strains, and curvature profiles. Comparisons of the ultimate

moment capacities, measured displacement ductilities, plastic hinge lengths,

cyclic energy absorption capacity and the failure modes of the full-scale

specimens are made with those observed from 1/6-scale model tests.

Keywords: Axial Load; Inelastic Behavior; Bridges; Columns; Concrete;

Confinement; Cyclic Load; Ductility; Energy Absorption; Failure; Full Scale;

Lateral Load; Microconcrete; Modeling; Plastic Hinge; Scale Effects.

in
ACKNOWLEDGEMENT

The authors would like to extend their sincere thanks Mr. Jim Little, Mr..

Frank Rankin, Mr. Erik Anderson, and Mr. Herb Wechsler of the Center for

Building Technology laboratory staff for their help in conducting both the

full-scale and model tests. The assistance Mr. Stuart Foltz, Mr. Richard

Morales, Dr. John Gross, Dr. Charles Scribner, Mr. Charles Yancey, and Mr

Mark Johnson of the CBT professional staff is also gratefully acknowledged.

These individuals often worked long hours and weekends to ensure the

successful conduct of the tests described in this report. The authors also

wish to extend their thanks to Mr. James H. Gates of the California Department

of Transportation, for serving as a continuing advisor on current design

practice for bridge columns in seismic regions, to Dr. Nigel Priestley of the

University of California, San Diego for his helpful comments on New Zealand

seismic design practice and current research projects at the University of

Canterbury, New Zealand and to Dr. William Gamble of the University of

Illinois, who assisted with the design of the full-scale test specimens and

reaction fixtures.

iv
PREFACE

The majority of highway and mass transit bridges in the United States with

reinforced concrete columns are older structures which were not designed for

earthquake loading or were designed with minimum criteria. Many of these

columns have failed in previous earthquakes. Dynamic, elastic anaylses of

column structures subjected to ground motions recorded during severe

earthquakes have shown that the resulting inertial loads are generally

i significantly greater than the static lateral loads recommended by previous

design codes. However, these structures can survive severe earthquakes if

they are ductile and can absorb and dissipate seismic energy by inelastic

action. This point was clearly demonstrated during both the 1971 San Fernando

earthquake as well as the more recent 1985 Mexico City earthquake, where

proper detailing often determined whether a structure survived or collapsed.

Energy dissipation provided by the development of ductile plastic hinges in

columns is essential to the satisfactory response under seismic loading of

many structures. A large portion of modern bridge structures constructed in

zones of high seismic activity are supported by piers consisting of one or

more columns. Inelastic response of these bridge structures under seismic

loads will inevitably involve the development of plastic hinges in the

columns, unless mechanical energy dissipators are incorporated in the design.

Behavior of bridge columns is thus fundamentally different from that of

building frames, where the strength design approach tries to ensure that

plastic hinges develop in beams. This is accomplished by specifying column

flexural and shear strengths to be higher than the maximum column loads that

will occur when hinges form in the beams

.

This basic difference in philosophy between building frames and

bridge frames means that much of the research on building frames is not

applicable to bridge seismic design. To date, only two countries,

New Zealand and Japan, have carried out extensive testing of bridge

columns to improve highway construction codes. There is still a paucity of

v
Figure I: Collapsed bridge in the San Fernando earthquake

vi
such research in the U.S. despite the problems made apparent in the 1971 San

Fernando earthquake, where 42 highway bridges received significant damage, and

five bridges collapsed [3] (see Figure I). Much of the damage was a consequence of inadequate column detailing resulting in:

a) insufficient ductility to accomodate the imposed displacements;

b) shear failure in shorter columns; and,

c) anchorage failure of longitudinal reinforcement in plastic hinges

forming at the column bases.

Damage to bridge piers in the San Fernando earthquake highlighted the need for

reassessment of existing seismic design practice for bridges in the U.S.

Since 1971, column design requirements have been modified. They now require

additional confinement steel to avoid "birdcaging" or compression buckling of

the longitudinal steel (see Figure II), and continuous steel at the footings

and pier cap to avoid the pull-out problem (see Figure III) . Until the

initiation of the test program described in this report, the adequacy of these

new designs had not been verified.

It is now widely accepted that adequate ductility of column plastic hinges can

only be obtained if sufficient transverse reinforcement is provided to confine

the concrete core of the column, to prevent lateral buckling of the

longitudinal flexural reinforcement, and to provide adequate shear

reinforcement. During the San Fernando earthquake, the failure of the columns

of several bridges and buildings could be attributed directly to inadequate

confinement of the plastic hinge regions. However, the amount and

distribution of confining reinforcement necessary to insure adequate ductility

without significant strength degradation is still a matter of controversy.

It is important to note that nearly all present design codes for bridge column

seismic details (with the previously noted exceptions of New Zealand and

Japan) are based on results of extensive research programs for building

columns. Building columns are generally much smaller in cross section (12- 15

in) than bridge columns which are typically 48-60 inches or larger in crosssection. Reinforcement ratios are also very different for bridge columns.

vii
These differences and others, which are discussed in Reference 1, may lead to

substantially different performance of a bridge column in a seismic event.

In a workshop on earthquake resistance of highway bridges in 1979, the Applied

Technology Council (ATC) stated that,

"There is a pressing need for experimental studies to determine the reserve

capacities of various bridge components. Much of the considerable research

work on column behavior has been done on relatively small specimens and has

been extrapolated for bridges from tests of columns typically used in

buildings. Bridge columns are larger and (usually) lower stressed axially

than building columns and this does not permit easy extrapolation from the

present wealth of building column data. Therefore work is (urgently)

needed to determine whether the behavior of small sections can be

extrapolated to larger cross sections [2]."

Furthermore, the ASCE-TCLEE Task Committee on research needs stated in March

1979 that,

"experimental testing of selected reinforced concrete (bridge) columns

should be performed to determine the lateral resistance and adequacy of

reinforcement. Particular emphasis should be placed on those columns

designed using pre-1971 California criteria."

Based on these recommendations the National Bureau of Standards proposed, in

the fall of 1980, a test program to be known as the "Large Scale Bridge

Column Project." Due to the high costs associated with performing full scale

tests, sufficient funding did not become available until mid- 1983. At this

time, design work began on the specimens -- full scale, 60 -inch (1.52 m)

diameter columns -- as well as the necessary laboratory test fixtures.

Sponsors for the project included the National Science Foundation (NSF) , the

National Bureau of Standards (NBS) , the Federal Highway Administration (FHWA)

and the California Department of Transportation (CALTRANS) . The objectives

of the project were to address the following topics:

viii
Figure II: Longitudinal bar buckling - "Bird-caging effect"

IX
Figure III: Pullout of longitudinal bars from the foundation

x
a) The effect of scale factor on bridge column design (i.e. could models

be effectively used to predict full-scale behavior)

b) The effectiveness of current design details (i.e. determine ductility

capacity of columns)

.

c) Identification of symptomatic problems in present detailing

practices

.

The project was initially divided into three phases. The first phase

consisted of constructing and testing two full scale, 60- inch (1.52 m)

diameter, spirally confined bridge columns, designed to recent CALTRANS

specifications. These specimens were to serve as benchmark data for

subsequent model tests and to verify at full scale the performance of the

post-1971 design requirements. The prototype specimens were to replicate, as

closely as possible, actual bridge columns and the boundary and loading

conditions that would be experienced by a bridge column in the field. A

minimum of two prototype tests were to be conducted to investigate two general

classes of bridge columns currently in use in seismically active regions of

the U.S. These included a short 15 foot (4,6 m) high column which would be

susceptible to shear type failures and a tall 30 foot (9.2 m) high column

which would be used to investigate the performance of a predominately

flexure -type column with continuous longitudinal reinforcement through the

plastic hinge region. A special computer-controlled testing laboratory, known

as the NBS Large Scale Structural Research Facility, was designed and

constructed to apply column axial loads of 12,000 kip (53.4 MN) which

simulated dead weight of the bridge superstructure and lateral loads of up to

1,200 kip (5.34 MN) which caused column moments of up to 54,000 kip- feet (73.3

MN-m) . Specimens weighing up to 4800 kip (21.36 MN) with heights of up to 60

feet (18.5 m) and column diameters of up to 8 feet (2.44 m) could be

accomodated in the facility with access from a casting yard by means of a rail

transport system.

Phase II, which was conducted in parallel with Phase I, involved the

construction and testing of 1/6 -scale models of the prototypes under identical

load histories and boundary conditions. Data gathering and sensor layout for

the model specimens were designed to be identical to those of the prototype so

that a direct comparison could be made. A further variable studied in this

phase was the effect of using microconcrete, which is the current recommended

xi
structural modelling practice [23], versus the use of a small nominal maximum

size aggegate ready-mix concrete. The chief advantage of the latter was that

it was less expensive.

The third and final phase of the project involved making comparisons between

the model and prototype specimens. Such comparisons were to be based on

ductility factor, energy absorption capacity, ultimate moment capacity,

plastic hinge length, and extent of yield penetration in the longitudinal

reinforcement

.

This report is the second in a series of reports which discusses the results

of the NBS Large Scale Bridge Column Project. This report presents the

findings of Phase I and III described above. A detailed description of Phase

II (model tests) as well as a literature review of previous bridge column

research is presented in Reference 1.

xii
Table of Contents

Abstract iii

Acknowledgement i-v

Preface v

Table of Contents xiii

1.0 Introduction 1

1.1 General 1

1.2 Objectives 2

1.3 Test Outline ' 3

2.0 Design of Test Specimens 9

3.0 Test Results and Observations 15

3 . 1 Introduction 15

3.2 Flexure Specimen 17

3.2.1 Material Properties 17

3.2.2 Ductility Factor (DF) - 1, Cycle 1 . 17

3.2.3 Ductility Factor - 2, Cycles 2 & 3 19

3.2.4 Ductility Factor - 3, Cycles 4 & 5 20

3.2.5 Ductility Factor - 4, Cycles 6-15 26

3.2.6 Ductility Factor - 5, Cycles 16-18 33

3.2.7 Ductility Factor - 6, Cycles 19-21 33

3.2.8 Ductility Factor - 6.6, Cycle 22 36

3.3 Shear Specimen 42

3.3.1 Material Properties 42

3.3.2 Ductility Factor - 1, Cycle 1 .... 42

3.3.3 Ductility Factor — 2, Cycles 2-3 . 46

3.3.4 Ductility Factor - 4, Cycles 4-5 . 46

3.3.5 Ductility Factor - 6, Cycles 6-7 . 53

3.3.6 Ductility Factor - 8, Cycles 8-9 . 58

3.3.7 Ductility Factor - 10, Cycles 10-12 58

4.0 Discussion of Results 67

4.1 Flexure Column Moments and Displacements 67

4.2 Shear Column Moments and Displacements .. 84

4.3 Structural Ductility: Comparison of Test

Results with New Zealand Design Charts

for Structural Ductility 96

4.4 Strain in Longitudinal Reinforcement .... 99

4.5 Strains in Confining Reinforcement ...... 109

4.6 Column Curvatures and Estimation of

xiii
Plastic Hinge Lengths 109

4.7 Inelastic Strain Energy Capacity . 125

4.8 Similitude Relationships: Model

versus Prototype 125

5.0 Conclusions 143

Appendix A 147

A.l Laboratory Related Problems and Logistics 147

A. 1.1 Introduction 147

A. 1.2 The Moving Specimen Concept 154

A. 1.3 Rail Transport System 159

A. 1.4 Reaction Wall System 167

A. 1.5 Deflection Considerations 172

A. 1.6 Tie-Back Frame 175

A. 1.7 Full-Scale Hinge Boundary Condition 180

A. 1.8 Apparatus Connections 183

A. 1.9 Loading Systems 188

A. 2 Mechanics of the Test Apparatus 192

A. 3 Access Techniques for Large Scale Tests.. 195

A.

4

Specimen Demolition and Removal 200

A. 5 Material and Construction Process ....... 203

A. 5.1 Material 203

A. 5. 2 Construction Process 205

A. 6 Instrumentation 209

A. 6. 2 Clip Gages 214

A. 6.

2

Flatcells and the Measurement

of Base Reaction Forces 216

A. 6.

3

Tie-Back Frame as a Load Ceil ..... 226

Appendix B 231

Appendix C 243

Appendix D 247

References 251

xiv
1 . 0 INTRODUCTION

1 . 1 General

Full scale testing of a structure or structural component is generally called

for when there is a lack of empirical data validating the procedure of

extrapolating full scale behavior from that of model specimens. Testing of

full scale prototypes is a costly and time consuming proposition, particularly

in terms of instrumentation and construction costs, which may include

fabrication of unusual test apparatus for use on only a limited number of

tests. With this constraint, design guidelines for bridge columns have

primarily relied upon the results obtained from model tests. Furthermore, the

design guidelines for bridge columns in the United States have been based

largely on tests of model building columns. The design differences between

building columns and bridge columns are discussed in Reference 1.

As a result of severe damage sustained by many bridge structures following the

1971 San Fernando earthquake, substantial modifications were made to the

seismic design guidelines for the state of California [4,5]. The need for

experimental testing of these new procedures, at full-scale, was first

proposed at the Applied Technology Council Workshop on Earthquake Resistance

of Highway Bridges held in the spring of 1979 [2] . In response to the

findings of Reference 2, an experimental test program was initiated by the

National Bureau of Standards in the spring of 1983 to investigate the seismic

performance of full-scale bridge columns. The test program was sponsored by

the National Science Foundation (NSF) , the California Department of

Transportation (CALTRANS) , the Federal Highway Administration (FHWA) and the

National Bureau of Standards (NBS) . The program involved the testing of full

scale and replicate model bridge columns in an effort to determine the effects

of scale and the effectiveness of the new reinforcement details adopted by

CALTRANS and AASHTO. This constitutes, to the authors knowledge, the first

laboratory test of full scale bridge columns subjected to quasi-static cyclic

lateral loading with combined axial load.

1
This report is the second in a series describing the NBS Bridge Column Test

Program. The results of six 1/6 scale model tests (the actual length scale

factor, s^, was 6.1), a literature survey, and background information on the

test program are presented in Reference 1. To perform the tests described in

this report many engineering problems had to be solved, particularly in the

construction of the testing apparatus for the full-scale specimens and in the

measures necessary to ensure similitude between model and prototype. A

discussion of the techniques used and their effectiveness, as well as

alternative approaches which were considered during design of test apparatus,

is presented in Appendix A.

Two full-scale tests were carried out. Because specimen geometry was

different for the two tests, the data presented should be carefully considered

with the understanding that each test represents an initial benchmark for

comparison and that further full scale testing of replicate specimens is

merited in order to establish a statistical basis for the relationships and

behavior described. Some of the results presented in Chapters 4 and 5

indicate that alternative specimen construction techniques are possible which

may result in a substantial reduction in the cost of testing. This will make

future full-scale testing more economical. Recommendations for future fullscale test procedures are presented in Appendix B.

1.2 Objectives

It was realized at the beginning that only a very limited number of full scale

specimens could be tested. Three were originally considered desireable, but

budget and manpower limitations eventually caused the number to be reduced to

two. Specific objectives were as follows:

1. To determine the ductility capacity of representative bridge columns

designed in accordance with recent CALTRANS standards.

2. To determine the effects of scale on column behavior.

3. To determine the effects of aspect ratio (L/D) on column behavior.

2
4. To determine the adequacy of transverse reinforcement as presently

required by CALTRANS for prevention of longitudinal bar buckling.

The two cantilevered bridge columns described in this report were designed and

constructed in accordance with recent CALTRANS specifications [4] . These

specifications meet or exceed the "Seismic Design Guidelines for Highway

Bridges" [5]. One column had an aspect ratio (Height / Diameter) of 6; the

other 3. These aspect ratios were chosen so that both flexure- dominated and

shear -dominated failures could be studied. Evaluation of column performance

was based on the following criteria:

1. Energy absorption capacity: a measure of a column's ability

to dissipate seismic energy through inelastic deformation.

This can be either the area bounded by a single load-deflection

hysteresis cycle, or the summation of all such cycles up to

failure of the column. This presently appears to be the most

effective measure of seismic performance.

2. Displacement ductility factor: maximum lateral deflection

at column failure divided by the yield deflection. This is a relative

measure of energy absorption capacity now used in some code

specifications

.

3. Ultimate moment capacity: maximum moment occurring prior to column

failure. Useful for comparison with analytically predicted capacity.

4. Effectiveness of transverse confinement: qualitative observations

concerning the ability of present confining reinforcement requirements

to prevent buckling of longitudinal steel and extend ductility.

1.3 Test Outline

The dimensions of the two full scale columns are presented in Figs. 1.1, 1.2

and Table 1.1 and are typical of current design practice. The height of the

flexure column, which had an aspect ratio of 6, was 30 ft. (9.14 m) measured*

from the top of the footing to the hinge at the top of the. column. The

3
4

Figure

1.1

Side

Profile,

Full-Scale

and

Model

Column
©Oo ©

T3

coo

CO '5co

CM <I <Coo

0) cn

©o©

"5

LL

f= 5

Figure

1.2

Column

Cross

Sections

and

Reinfor<
The height of the shear column, which had an aspect ratio of 3, was 15 ft.

(4.57 m) . Both columns were circular in cross section with a diameter of 5

ft. (1.52 m) . The longitudinal reinforcement consisted of 25 - #14 (1.69 in.,

43 mm) deformed steel reinforcing bars, grade 60 (60 ksi nominal yield),

spaced evenly around the perimeter of the column. The transverse

reinforcement consisted of #5 (0.625 in., 16 mm), grade 60 deformed steel

spirals with a pitch of 3 - 1/2 in. (88.9 mm) for the flexure column and #6

(0.75 in., 19 mm), grade 60 deformed steel spirals with a pitch of 2 - 1/8 in.

(53.97 mm) for the shear column. Material specifications are presented in

Appendix A, Tables A.l thru A. 3.

The boundary conditions for the columns were, for all test specimens, fixed at

the column base and hinged at the top. These were felt to best model the insitu boundary conditions for a broad class of bridge columns used in seismic

regions, i.e. those which can be assumed to act as single cantilevered

sections with the bridge superstructure supported on elastomeric bearing pads.

Furthermore an axial column load which simulated the vertical reaction of the

bridge superstructure was to be applied in such a manner as to maintain a

vertical reference axis throughout the test. This precluded the use of axial

post- tensioning for application of the vertical force, since that method does

not produce the P- delta effects experienced in the real structure under the

influence of lateral loads . The mechanisms by which these boundary conditions

were achieved in the laboratory for the model and prototype specimens were

quite different and are discussed in detail in Chapter 2.

The columns were first loaded axially to a specified load of 1000 kip (4,448.2

kN) , a value equal to approximately 0.1 f' c Ag. Axial load was held constant

during the test and was the same for both specimens. The columns were then

subjected to reversed cyclic lateral loads. The yield deflection (referred to

as Ay throughout this report) was determined experimentally as described in

Chapter 3. Thereafter, cyclic lateral deflections were increased by multiples

of Ay. The lateral load was applied at the center of gravity of the column

footing. Loading histories for the columns are discussed in Chapter 3. The

cycling of the columns was performed under displacement control using a double

acting servo -hydraulic jack. Axial loads were applied under load control

6
under load control using a specially converted servo -hydraulic universal

testing machine. Strains in spirals and along the longitudinal bars were

measured using strain gages. External displacement transducers were used to

measure column rotation, lateral displacement and column curvature. All

passive test apparatus (e.g. supports, lateral restraint systems) were

instrumented and calibrated for monitoring reactions because the specimen and

test apparatus constituted a statically indeterminant system.

Table 1.1 Column Dimensions

Specimen Type Height Diameter Axial

of ( inches) Load

Concrete (kip)

Model Columns

N1 Microconcrete V - 5.5" 9.8 26.87

N2 Microconcrete 2

- 5.5" 9.8 53.75

N3 Microconcrete 4' - 11" 9.8 26.87

N4 Ready -Mix

(1/4" Gravel)

2

- 5.5" 9.8 26.87

N5 Ready -Mix

(1/4" Gravel)

2' - 5.5" 9.8 53.75

N6 Ready -Mix

(1/4" Gravel)

4' - 11" 9.8 26.87

Prototype Columns

Flexure Ready -Mix

(3/4" Gravel)

30' - 0" 60.0 1000.00

Shear Ready -Mix 15' - 0" 60.0 1000.00

(3/4" Gravel)

7
'

.

s
2.0 DESIGN OF TEST SPECIMENS

Both columns were built to meet CALTRANS specifications effective in 1983 [4] . The final column diameter, aspect ratio, and reinforcing details were selected

on the basis of discussions with CALTRANS engineers [18] to reflect standard

field practice and representative column sizes. The resulting dimensions were

a height of 30 ft. (9.14 m) for the flexure column and 15 ft. (4.57 m) for the

shear column with a diameter of 5 ft. (1.52 m) for both columns.

The longitudinal reinforcement for both the flexure and shear columns

consisted of 25 continuous lengths of grade 60 #14 deformed bars (1.69 in., 43

mm). This resulted in a longitudinal reinforcement ratio, p t , of 2%. The

center- to-center spacing of the longitudinal bars was 6.83 in. (173 mm).

CALTRANS specifies that neither welded nor lapped splices are permitted for

longitudinal reinforcement larger than #11 bars (1.41 in., 35.8 mm) within a region extending 10 ft. (3.0 m) above the column footing for a fixed base

column.

The transverse reinforcement for the flexure column was a spiral made from #5

(0.625 in., 15.9 mm) grade 60 deformed steel bars with a pitch of 3.5 in.

(88.9 mm) resulting in a volumetric reinforcement ratio, p s , of 0.00633. The

spiral for the shear column was made from #6 (0.75 in., 19.1 mm) grade 60

deformed steel bars spaced at 2 - 1/8 in. (54.0 mm) giving a volumetric

reinforcement ratio of 0.01479. The spiral extended from the top of the

column into • the column footing to the point of tangency of the longitudinal

bar hooks as per CALTRANS Design Handbook [6] Section 8.18.2.1.3. The

dimensions and steel arrangement for both columns are shown in Figs. 2.1-

2.3.

CALTRANS specifies a minimum amount of transverse reinforcement in their

design code for bridge columns [4]

.

The amount of transverse steel required

in the potential plastic hinge region is calculated from the following

equation for columns with diameters equal to or larger than 5 ft. (1.52 m):

9
10
SPECIMEN

END

PROFILE

11

Figure

2.2

Dimensions

and

Steel

Reinforcement

Detail

(Full-Scale

Column

Specimens,

Plan

and

End

Profile)
BOl

TOM

IONUIUUINAI

Ht

INKJHCfcMtN

!

12
P s > 0.12 ( 2 . 1 ) f

^ c pe

n s + i 9s

1

M-l 1f ' A C § . > 0.12 (5.2/60) [0.5 + 1.25 * (1000/5.2 * II * 30 2

) ]

> 0.00608 (based on the concrete compressive strength of

the prototype flexure column)

However, this reinforcement ratio must not be less than:

p s > 0.45 1

> 0.45 ( 30 2 /28 2 - 1) 5.2/60

> 0.00577

( 2 . 2 )

Equation 2.1, therefore, governed the amount of transverse reinforcement for

these columns. The potential plastic hinge length as defined by CALTPANS [4]

is the greater of:

1. The maximum horizontal dimension - 60 in. (1.52 m) .

2. 1/6 of the column length - 360/6 in. - 60 in. (1.52 m) Flexure

- 180/6 in. - 30 in. (0.74 m) Shear

3. 24 in. (0.61 m)

.

The potential plastic hinge length for both the flexure and shear columns is

therefore equal to 60 in. (1.52 m) .

The (Pe/f' c Ag) ratio for both columns was

p e/ f 'c A

g ” °- 09

based on a concrete design strength of 4000 psi (27.6 MPa) and an axial load

of 1000 kip (4,448.2 kN) . Actual concrete strengths obtained from compression

13
tests of 12 x 6 in. (305 x 122 mm) cylinders were 5200 psi (35.9 MPa) for the

flexure column and 4980 psi (28.54 MPa) for the shear column. These led to

the following axial loading conditions for the two specimens:

Pe/f'c Ag - 0.068

Pe/f'c Ag - 0.085

[flexure specimen]

[shear specimen]

The following chapter describes the results of the prototype tests. The means

by which the tests were carried out are described in detail in Appendix A.

14
3.0 TEST RESULTS AND OBSERVATIONS

3 . 1 Introduction

The method used to determine the yield displacement was the same as that used

for the model tests [1] . In brief, the ultimate moment of the column was

determined using ACI column charts [11]. From this moment value, an ultimate

lateral load was obtained. The columns were then laterally loaded to 75% of

ultimate lateral load in the forward and then in the reverse direction.

Displacements were experimentally measured in both these directions. The

average of the two displacements divided by 0.75 was then used as the yield

displacement, Ay. This definition of the yield displacement was used by

Priestley et. al . [12] in their test programs at the University of Canterbury

and is graphically defined in Fig. 3.1.

While cyclic load test procedures for steel structures are in the process of

standardization [13], no such recommended procedures have yet been agreed upon

for testing reinforced concrete structures. The determination of Ay, and the

cyclic load history to which the structure will be subjected, is therefore

somewhat arbitrary. Unfortunately, there is evidence that column behavior is

dependent upon cyclic load history [14]. For these reasons, and because of

the wealth of bridge column test data available for comparison from New

Zealand, the procedure described above for the determination of yield

deflection, as well as the practice of using only two complete cycles at any

given level of displacement ductility, was adopted. Observations concerning

the validity of the method for the determination of yield displacement are

presented in Chapter 4.

The columns were tested in the NBS Large Scale Test Facility described in

Chapter 2. Loading history for each prototype column was identical to its

corresponding model column. The loading sequence for the flexure column was

one cycle to determine yield displacement, two cycles each at fj, equal to ±2

and +3, 10 cycles at /i equal to +4 and two cycles each at /i equal to +5, +6,

where /i, the displacement ductility, is defined as the ratio

15
Figure 3.1 Experimental Definition of Yield Deflection

as per Reference 12.

16
of the displacement at any point to the yield displacement. A third cycle

was conducted if at a given ductility level, the moment capacity of the

column in the second cycle was significantly lower than that obtained in the

first cycle.

The columns were first loaded axially to 1,000 kip (4,448.2 kN) . This load

was chosen as being representative of the gravity load due to the bridge

superstructure. The columns were then laterally loaded in the east-west

direction with the first excursion to the east (forward direction) and then to

the west (reverse direction) . A complete cycle was defined as an excursion to

the east followed by a return to the initial position and a subsequent

excursion to the west followed by a return to the initial position. The

entire test was performed under displacement control.

A detailed discussion of the observations and behavior of the columns during

each test is presented in this chapter. The discussion is presented in a

cycle -by- cycle format augmented by figures to depict any significant changes

in column appearance. A discussion of results is presented in Chapter 4.

3 . 2 Flexure Specimen

3.2.1 MATERIAL PROPERTIES

f' c - 5200 psi (35.8 MPa)

Pe - 1000 kip (4,448.2 kN)

Mu(experimental) - 9643 kip-ft (13 MN-m)

Ph - 200 kip (889.6 kN)

Ay (experimental) - 3.53 in. (89.7 mm)

Where Pe is the axial load imposed on the column, P^ is lateral load equal to

75% of the ultimate lateral load, and Mu

is the maximum moment achieved

experimentally. The loading history for the column is shown in Fig. 3.2

3.2.2 DUCTILITY FACTOR (DF) - 1, CYCLE 1

Flexural cracking spaced at approximately 1 ft. (305 mm) apart was observed on

both the east and west sides of the column. Cracking on the west side was

first noted at a lateral load of 36.8 kip and at a height of 4 - 1/2 ft.

17
LOADING

SEQUENCE

FOR

FULL-SCALE

FLEXURE

COLUMN

d d

18

Figure

3.2

Loading

History

for

Full-Scale

Flexure

Column
(1372 mm) above the base of the column. The length of the cracks were

approximately 4 ft. (1219 mm) long. Upon loading to 200 kip (389.6 kN)

laterally to the west, flexural cracks were observed up to approximately 10

ft. (3048 mm) and some of the existing cracks had propagated to the middle of

the column i.e. the north- south axis. The measured displacement upon

loading laterally to 200 kip to the east was 2.500 in. (63.5 mm) and 2.799 in.

(71 mm) to the west. The yield displacement for this column was therefore

equal to 3.533 in. (89.7 mm).

Ay - , 4 (2.5 + 2.799) “ 3

- 3.533 in.

See Section 3.1 for an explanation of the method used to calculate yield

displacement

.

Because of the instrumentation difficulties described in Chapter 2 regarding

the direct measurement of friction forces developed in the roller support

system, the actual lateral load seen by the column (P-^) was somewhat less than

200 kip (889.6 kN) . It will be shown in Chapter 4 that these friction forces

resulted in the underestimation of Ay by approximataly 2% for the full scale

Flexure specimen and 1% for the full scale Shear specimen. This factor is

taken into account in later comparisons between prototype and model specimens.

3.2.3 DUCTILITY FACTOR - 2, CYCLES 2 & 3

Cycle 2: Extensive formation of additional cracks were noted. Vertical

cracks were observed forming between the flexural cracks up to a height of

approximately 8 ft. (2438 mm) . Some diagonal cracks (approximately 45°) were

noted on the north and south side of the column. Propagation of existing

flexural cracks to the north- south axis at a height of 10 ft. (3048 mm)

occurred. Incipient spalling at the base of the column on the west side was

noted. No spalling on the east side was observed. Again, cracking of

concrete could be heard. Cracks on the top face of the footing on the east

side developed. Flexure cracks in the footing on the east and north faces

19
occurred. These cracks occurred at the what appeared to the pour line. The

crack pattern on the west side of the column is shown in Fig. 3.3.

Cycle 3: Upon reaching 2 Ay to the east, 2 pieces of concrete cover on the

west side spalled off.

3.2.4 DUCTILITY FACTOR - 3, CYCLES 4 & 5

Cycle 4: Flaking of concrete on the west side of the column up to a height of

4 ft. -10 in. (1473 mm) was noted on loading to the east. Fig. 3.4 shows this

flaking of the west side of the column. The maximum width of the flake area

was 3 ft. -10 in. (1168 mm). Additional flexural and shear cracks, mainly in

the bottom third of the column, were observed.

On loading to the west, a piece of concrete cover 1 ft. - 7 in. (L) x 2 ft.-

1 in. (W) [483 mm x 635 mm] spalled off on the west side of the column as

shown in Fig. 3.5. The extent of cracking on the west side of the column can

also be clearly seen in Fig. 3.6. The east face began to show signs of

flaking. The area of flaking was approximately 2 ft. x 3 ft. [610 mm x 914

mm] (Note all dimensions are width x height) . Some additional flexural and

shear cracks were noted. Two additional cracks on the top face of the footing

on the west side were noted.

Cycle 5: Several pieces of cover concrete spalled off on the east side of the

column on the excursion to the east. The spall area was 1 ft. -10 in. x 2 ft.

- 1 in. (559 mm x 635 mm) and is shown in Fig. 3.7. The depth of the cover

concrete was approximately 4 in. (102 mm) as compared with a specified cover

of 2 in. (51 mm) in the plan drawings. Additional spalling of the west side

of the column occurred increasing the spall area to approximately 2 ft. x 3

ft. - 2 in. (610 x 965 mm). The spalling of the concrete cover did not,

however, expose any reinforcing steel at this stage of the test. A shear

crack on the north face and another on the south face of the footing

coinciding with the north- south axis of the column were noted.

20
Figure 3.3: Flexure Specimen

Displacement - 2Ay; Cycle 1 East

View of West Side of Column

21
Figure 3.4: Flexure Specimen

Displacement * 3Ay; Cycle 1 East

View of East Side of Column

22
Figure 3.5: Flexure Specimen

Displacement - 3Ay; Cycle 1 West;

View of West Side of Column.

Note Technician Marking Cracks on

Right Side of Column.

23
Figure 3.6: Flexure Specimen

Displacement * 3Ay; Cycle 2 West

View of West Side of Column Showing

Commencement of Spalling

24
Figure 3.7: Flexure Specimen

Displacement - 4Ay; Cycle 1 East

View of West Side of Column

25
3.2.5 DUCTILITY FACTOR - 4, CYCLES 6 - 15

Cycle 6: Shear cracks were noted on the north and south faces of the footing.

Additional spalling of the west side of the column took place exposing the

spiral on that side. Views of the west side of the column before spalling and

after spalling are shown in Figs. 3.8 and 3.9. The exposed spiral on the west

side is indicated on Fig. 3.9. The spall area was approximately 1 ft. - 7 in.

x 3 ft. - 6 in. (48.3 cm x 106.7 cm) on the west side of the column and 3 ft.

<• x 2 ft. (91.4 cm x 61 cm) on the east side of the column. The spall area on

i the east side is shown in Fig. 3.10. The thickness of the cover concrete on

the west side was also approximately 4 in. (10.2 cm). Some additional shear

cracks formed on the north side of the column mainly in the bottom third of

the column.

Cycle 7: Crackwidths on the east side of the column were approximately 1/4

in. (6.35 mm) at a height of 3 ft. - 4 in. (101.6 cm) and at 3 ft. - 9 in.

(114.3 cm) above the column base. The spiral was still not exposed on the

east side of the column at this stage although additional spalling of the

cover concrete occurred in this cycle. Fig. 3.11 shows the east face of the

column and the extent of the spall area.

Cycles 8 - 15: Additional spalling of the cover concrete took place over

these eight cycles with no significant changes to the spall areas or crack

pattern occurring. The spiral on the east side of the column was exposed

during the 12^ cycle. Figs. 3.12 and 3.13 show the spall areas on the east

and west sides of the column respectively. The LVDT measuring the rotation on

the west side was removed due to the loss of the cover concrete which held it

in place.

26
JU

:

.

- -

Figure 3.8: Flexure Specimen

Displacement - 4Ay; Cycle 1 West

View of West Side of Column

27
Figure 3.9: Flexure Specimen

Displacement — 4Ay; Cycle 1 East

View of East Side of Column

28
Figure 3.10: Flexure Specimen

Displacement - 4Ay; Cycle 2 West

View of East Side of Column

29
Figure 3.11: Flexure Specimen

Displacement - 4Ay; Cycle 10 West

View of East Side of Column

30
Figure 3.12: Flexure Specimen

Displacement - 4Ay; Cycle 10 East

View of West Side of Column

31
Figure 3.13: Flexure Specimen

Displacement - 5Ay; Cycle 3 West

View of West Side of Column

32
3.2.6 DUCTILITY FACTOR - 5, CYCLES 16 - 18

Cycles 16 - 18: Approximately 7 shear cracks were noted on the south face of

the footing and about 2 were noted on the north face of the footing. Again,

only minor spalling occurred with no significant changes in the crack pattern

or spall areas during these three cycles. The spall areas on the east side of

the column was 4 ft. - 6 in. x 3 ft. - 6 1/2 in. (1372 mm x 1079 mm) and 4 ft.

- 3 in. x 4 ft. - 2 in. (1295 mm x 1270 mm) on the west side. The crackwidth

on the west side was approximately 3/8 in. (95 mm) at a height of 3 ft. - 5

in. (1041 mm) above the top of the footing. Figs. 3.14 and 3.15 show the

spall area on the west and east side of the column with the main reinforcing

bars exposed, respectively.

A third cycle, cycle 18, was completed at this ductility level not because of

a degradation in the load carrying capcity of the column was noted but because

the corresponding model, model N6 , was cycled three times at this ductility

level

.

3.2.7 DUCTITLITY FACTOR - 6, CYCLES 19 - 21

Cycle 19: The two longitudinal bars on the center of the west side of the

column were noted to have bowed out by approximately 1 in. (25.4 mm). The

spiral between these two bars showed signs of straightening which indicates

yielding.

Cycle 20: A spiral on the west side of the column at a height of 1 ft. - 5

1/4 in. (438 mm) fractured during the excursion to the east. The fracture

occurred where the spiral was spliced. Fig. 3.16 shows the fractured spiral

on the left side of the photo and the bowed longitudinal bars on the right.

During the excursion to the west, a longitudinal bar fractured on the west

side of the column. A drop in lateral load of approximately 13% corresponded

to the fractures of the spiral and the longitudinal bar.

33
34

Figure

3.14:

Full

Scale

Flexure

Specimen

Figure

3.15:

Full

Scale

Flexure

Specimen

Displacement

=

5Ay;

Cycle

2

West

Displacement

=

6Ay;

Cycle

2

East

View

of

East

Side

of

Column

View

of

West

side

of

Column
Figure 3.16: Flexure Specimen

Displacement - 6Ay; Cycle 3 East

View of West Side of Column

35
Cycle 21: On the excursion east, a second spiral on the west side below the

first fractured spiral fractured. A third spiral located right above the

first fractured spiral fractured shortly after the fracture of the second

spiral. The three fractured spirals in addition to the longitudinal bar which

fractured in the last cycle are shown in Fig. 3.17. During the excursion

west, three additional longitudinal bars on the west side of the column

fractured. On the subsequent excursion back to initial' position, a fourth

spiral fractured on the west side. The fractured longitudinal bars are shown

in Fig . 3.18.

3.2.8 DUCTILITY FACTOR - 6.6, CYCLE 22

Due to the limitation of 23.4 in. (594 mm) set by the maximum stroke of the

servo -hydraulic ram, a maximum ductility factor achievable for this test was

6.6. This did not, however, pose a problem as the column had already reached

its failure state at the previous ductility level as will be discussed in

further detail in Chapter 4.

Two vertical bars on the east side of the column fractured during the

excursion to the east as shown in Fig. 3.19. Fig. 3.20 shows the west side of

the column during the excursion east. On the excursion west, two additional

longitudinal bars on the west side fractured. This made a total of six

fractured longitudinal bars on the west side and two on the east side. The

six fractured bars on the west is graphically shown in Fig. 3.20. The extent

of the lateral displacement to the column is shown in Fig. 3.21.

36
Figure 3.17: Flexure Specimen

Displacement - 6Ay; Cycle 3 West

View of West Side of Column

37
Figure 3.18: Flexure Specimen

Displacement =* 7Ay; Cycle 1 East

View of East Side of Column

38
Figure 3.19: Flexure Specimen

Displacement = 7 Ay; Cycle 1 East

View of West Side of Column

39
Figure 3.20: Flexure Specimen

Displacement - 7 Ay; Cycle 1 West

View of West Side of Column

40
Figure 3.21: Flexure Specimen

Displacement = 7Ay; Cycle 1 East

Overall View of Test

41
3.3 Shear Specimen

3.3.1 MATERIAL PROPERTIES

f' c

=. 4,140 psi (28.5 MPa)

P e = 1000 kip (4,448.2 kN)

M^ (experimental) = 11640 (15.8 MN-m)

Ph

= 400 kip (1779.3 kN)

Ay (experimental) = 1.40 in. (35.6 mm)

i

The loading history for the shear specimen is shown in Fig. 3.22. Note that

the first excursion was to the west instead of to the east as with the flexure

column

.

3.3.2 DUCITILITY FACTOR = 1, CYCLE 1

The first flexural crack, 2 ft. above the base, was noted on the west side at

a lateral load of 191.5 kip on the excursion west. Hairline flexural cracks

running from north to south were observed on the west side up to a height of 8

ft. 10 in. (2692 mm). A crack at the column/base joint on the west side was

noted. Hairline flexural cracks on the north and south faces of the column

footing at approximately mid-height of the footing were also noted. Fig. 3.23

shows a view of the west side of the column.

First cracking of the east side of the column occurred at a lateral load of

140 kip on the excursion east. Only flexural cracks were noted on the east

side of the column during this excursion up to a height of 9 ft. 3 in. (2819

mm). The crack pattern is shown in Fig. 3.24.

The measured displacements at lateral load equal to 400 kip were 1.144 in.

(29.1 mm) on the excursion west and 0.951 in. (24.2 mm) on the excursion east.

The yield displacement was therefore equal to 1.40 in. (35.6 mm).

Av = 4 (1.144 + 0.951)

y in.

3

= 1.40 in.

42
jX

Figure 3.22 Loading History for Shear Column

43
Figure 3.23: Shear Specimen

Displacement = lAy; Cycle 1 West

View of West Side of Column

44

1
Figure 3.24: Shear Specimen

Displacement = lAy; Cycle 1 East

View of East Side of Column

45
Note that the affect of friction bewteen the hilman rollers and the track

system was not taken into account in the 400 kip (1779.3 kN) lateral load.

This was done so that this test would be consistent with the flexure test.

3.3.3 DUCTILITY FACTOR - 2, CYCLES 2 AND 3

Cycle 2: Extensive cracking was noted during this cycle. Flexural cracking

extended to the top of the column on both the east and west faces. Formation

of diagonal cracks at approximately 45° from horizontal on the north and south

faces were observed. The extent of the flexural cracks on the west face and

shear cracks on the north face are shown in Fig. 3.25.

Cycle 3: Some additional flexural and shear cracks were observed on the north

and south faces. However, no significant change in crack pattern was noted.

Fig. 3.26 shows a view of the east face of the column. An overall view of the

test set-up looking east is shown in Fig. 3.27.

3.3.4 DUCTILITY FACTOR - 4, CYCLES 4 AND 5

Cycle 4: Spalling of the east side of the column was noted at a lateral load

of approximately 578 kip. The spall area located on the southeast face was

approximately 10 in. x 23 in. (W x H) at a height of about 24 in. above the

column base. Maximum crackwidths on the west face was approximately 0.2 in.

On the excursion east, flexural cracks were noted on the east and west faces

in the upper third of the column. More shear cracking was also noted on the

north and south faces. Spalling of the west face of the column at

approximately 9 in. above the base also occurred during this excursion. The

spall area was approximately 4.5 ft. in the east-west direction. Some

additional spalling occurred on the east face. The spall areas on the east

and west faces are shown in Figs. 3.28 and 3.29 respectively. Fig. 3.30 shows

an overall view of the test set-up looking west.

46
Figure 3.25: Shear Specimen

Displacement = 2Ay; Cycle 1 West

View of West Side of Column

47
Figure 3.26: Shear Specimen

Displacement = 2Ay; Cycle 2 East

View of East Side of Column

48
Figure 3.27 Shear Specimen

Displacement = 2Ay; Cycle 2 East

Overall View of Test and West Side

of Column.

49
Figure 3.28: Shear Specimen

Displacement = 4Ay; Cycle 1 East

View of East Side of Column

50
Figure 3.29: Shear Specimen

Displacement = 4Ay; Cycle 1 East

View of West Side of Column

51
Figure 3.30: Shear Specimen

Displacement = 4Ay; Cycle 1 West

Overall View of East Side of Column

52
Cycle 5: No change in crack pattern was noted.

3.3.5 DUCTILITY FACTOR = 6, CYCLES 6 AND 7

Cycle 6: Spalling of the east and west faces of the column was noted.

Spirals at heights of 15, 10 and 7 (59.1, 39.4, 27.6 mm) inches above the top

of the footing were exposed on the west side due to the spalling. Spirals on

the east side were exposed at heights of 22, 20, 17, 14, 12, 8, and 6 (86.6,

78.7, 66.9, 55.1, 47.2, 31.5, and 23.6 mm) inches above the top of the

footing. The spalling disconnected some to the cables to the clip gages on

the east side.

The spall area on the west side was approximately 15 in. x 54 in. (381 mm x

1372 mm). The east face had spalled in 4 different locations. Fig. 3.31

shows the spall area on the west face of the column. Some additional shear

cracks were noted on the east and west faces of the column.

Extensive flexural cracking of the footing on the north and south faces was

noted. The crack patterns on these two faces were almost identical. Some of

these cracks were extensions of the radial cracks on the top of the footing.

Additional radial cracks on top of the footing on both the east and west sides

were also noted.

Cycle 7: Additional cracks of the footing were noted. No additional flexure

or shear cracks were noted. A significantly large piece of cover concrete on

the southeast side of the column was held in place only by the ties used in

construction of the steel cage. Fig. 3.32 shows the spall areas on the east

side and Fig. 3.33 shows the piece of concrete ready to spall off. The

cracking of the north face of the footing and the north face of the column is

shown in Fig. 3.34

53
Figure 3.31: Shear Specimen

Displacement = 6Ay; Cycle 1 East

View of West Side of Column

54
Figure 3.32: Shear Specimen

Displacement = 6Ay; Cycle 2 East

View of East Side of Column

55
Figure 3.33: Shear Specimen

Displacement = 6Ay; Cycle 2 East

View of West Side of Column

Showing Spalled Concrete Section.

56
Figure 3.34: Shear Specimen

Displacement - 6Ay; Cycle 2 East

View of North Side of Column

Showing Extent of Diagonal Cracking.

57
3.3.6 DUCTILITY FACTOR = 8 , CYCLES 8 AND 9

Cycle 8: On the excursion west, a muffled bang was heard. No visible signs

of spiral or vertical bar fracture were noted due to the concrete obscuring

most of the bars. However, the spirals on the west side of the column appear

to have buldged out. More shear cracks were noted on the north and south

faces of the column. Spalling of the east and west sides of the column

continued. Propagation of the radial cracks on the top of the footing and

cracks on the north face of the footing occurred. Figs. 3.35 and 3.36 show

the spall areas of the east and west side respectively.

Cycle 9: No significant change in crack pattern noted.

3.3.7 DUCTILITY FACTOR - 10, CYCLES 10,11 AND 12

Cycle 10: Some spalling occurred with no significant change in crack pattern

on the excursion west. Two loud bangs and a softer bang was heard on the

excursion east. These bangs appear to originate from the east side of the

column. Again, no visible evidence of whether the bangs were spiral or

vertical bars fracturing as the concrete core was still pretty much intact.

On the excursion back to center position from the east excursion, a vertical

bar on the west side of the column could to be seen to have fractured at a

height of 18 in. above the base. The spall area on the east side is

approximately 60 x 53 in. (1524 x 1346 mm) and the spall area on the west

side is approximately 32 x 53 in. (813 x 1346 mm). Some gauging of the

footing at the column-base joint was noted on the west side and is shown in

Fig. 3.37. The extensive spalling of the east face of the column is shown in

Fig. 3.38.

Cycle 11: No bangs were heard on the excursion west. On the excursion east,

several bangs were heard. A spiral on the northeast side at a height of 4 ft.

6 in. (1372 mm) fractured. The vertical bar directly behind the fractured

spiral also fractured. The location of the fracture was 5 ft. (1524 mm)

58
Figure 3.35: Shear Specimen

Displacement = 8Ay; Cycle 1 East

View of East Side of Column

59
Figure 3.36: Shear Specimen

Displacement = 8Ay; Cycle 1 East

View of West Side of Column

60
Figure 3.37: Shear Specimen

Displacement =* lOAy;

View of West Side of

Cycle 1

Column

East

61
Figure 3.38: Shear Specimen

Displacement = lOAy; Cycle 1 East

View of East Side of Column

62
above the base. A spiral on the west side approximately 8 in. (203 mm) above

the base could be seen to have fractured. Extensive buckling of the vertical

bars on the west was noted.

On the excursion back to center from the east, several bangs were heard. Two

vertical bars north of the east-west axis fractured at approximately 5 in.

(127 mm) above the base. A third bar on the west side fractured at a height

of 14 in. (356 mm) above the base. A total of 4 vertical bars and a spiral

could be seen to have fractured on the west and a total of one vertical bar

and a spiral fracture were visible on the east side. The buckled and

fractured bars on the west is shown in Fig. 3.39. An overall view of the

column is shown in Fig. 3.40.

Cycle 12: Two bangs which seemed to originate from the west side were heard

on the excursion west. However, no additional bar fractures were however

visible. The concrete core was observed to be breaking up.

On the excursion east, several loud bangs were heard from the east and west

sides. At the end of the eastward excursion, a total of 6 vertical bars and

one spiral hoop were observed to have fractured on the west side. A total of

5 vertical bars and a spiral on the east side were also fractured. All but

one of the longitudinal fractures occurred approximately 9 in. (229 mm) above

the base. The unusual bar failure was at a height of 5 ft. (1524 mm) above

the base approximately 20 degrees counterclockwise (second bar) from

centerline (maximum fiber distance from neutral axis) . This was a clean

fracture with no signs of prior yielding. On closer inspection a tack weld

was discovered on the inside face of the bar where a temporary support frame

had been attached during erection of the column reinforcement cage prior to

casting. Because this failure did not occur until after buckling- induced

fracture of the 11 longitudinal bars mentioned above it is felt that this had

no adverse affect on the test data. However, it does underscore the need to

avoid weldments, particularly involving reinforcing bar, in the potential

plastic hinge region.

63
On the return excursion to center, a soft bang was heard. At center position,

the following bars were noted to have fractured:

Bar location

(referenced to the

eastmost/ westmost

bar)

East

:

1

bar south

eastmost

1 bar north

2 bars north

3 bars north

4 bars north

Spiral, 4 bars north

West

:

3 bars south

2 bars south

1 bar south

westmost

1 bar north

2 bars north

3 bars north

Spiral, 2 bars north

Height of fracture, in. (mm)

above the base

11 (279)

6 (152)

8 (203)

9 (229)

53 (1346)

52 (1321)

48 (1219)

9 (229)

10 (254)

13 (330)

19 (483)

4 (102)

4 (102)

4 (102)

4 (102)

Final spall areas were approximately 5 ft. x 5 ft. (1524 x 1524 mm) on the

east side and 2 ft. 9 in. x 5 ft. 6 in. (838 x 1676 mm) on the west side.

64
Figure 3.39: Shear Specimen

Displacement = lOAy; Cycle 2 East

View of West Side of Column

65
Figure 3.40: Shear Specimen

Displacement = 10Ay; Cycle 2 West

Overall View of Test; West Side

is to Right of Column.

66
4.0 DISCUSSION OF RESULTS

There are several traditional methods of analyzing results of cyclic load

tests on reinforced concrete structures. These involve presentations of

load-deflection histories, longitudinal and confining steel strain profiles,

and comparisons of theoretical and experimental determinations of plastic

hinge lengths . All of these traditional methods are presented in this

chapter. The format that is used is the same as that used in references 7,

15, 16, and 17 in an effort to be compatible with previous work in the field.

The primary emphasis of this chapter, however, will be to address the two

principal questions posed at the beginning of the project:

1) How effective are recent CALTRANS bridge column detailing procedures

for resisting seismic loads?

2) Can full-scale bridge column behavior be predicted from model specimens?

Special attention is given to the study of cyclic inelastic strain energy

dissipation as a means of characterizing overall structural performance under

seismic loading. It will be shown later in this chapter that this appears to

provide the best measure of similitude between model and prototype behavior.

Tables 4.1 through 4.3 present details of the longitudinal and transverse

reinforcement, as well as calculated specimen strengths for both the fullscale and 1/6-scale model tests. Representative s tress - strain curves for the

various reinforcing steels are presented in Appendix B. Interaction diagrams

for both model and prototype specimens are presented in Figures 4.1 and 4.2.

The two different diagrams superimposed in Figure 4.1 resulted from differing

concrete strengths for the full-scale shear and flexure columns. However, for

axial loads significantly below the balance point both curves coincide and

thus both full-scale tests are represented by the same point.

4.1 Flexure Column Moments and Displacements

As described in Chapter 2 there were two approaches possible for the

determination of column base moment and shear. Due to the early failure of

the flatcell electronics during the full scale flexure test and a power outage

67
Table 4 .

1

Details of the Transverse Reinforcement and Concrete for

Full-Scale and 1/6-Scale Model Specimens.

Plain Cone

.

Transverse Reinforcing Confined Concrete

Specimen

f

L c (psi)

Ec

(ksi)

Bar Dia.

(in. ) Spacing(in. ) fs fyh

(ksi)

fi

(ksi)

f

L cc

(psi)

€ max

N1 3.49 3367 0.120 0.350 .0145 64.0 .454 5.40 .023

N2 3.35 3299 0.120 0.350 .0145 64.0 .454 5.25 .023

N4 3.54 3391 0.120 0.350 .0145 64.0 .454 5.45 .023

N5 3.53 3387 0.120 0.350 .0145 64.0 .454 5.45 .023

SHEAR 4.98 4022 0.750 2.125 .0152 *63.1 .457 5.93 .024

N3 3.68 3458 0.106 0.570 .0069 69.0 .228 4.66 .010

N6 3.38 3352 0.106 0.570 .0069 69.0 .228 4.35 .010

FLEX 5.20 4110 0.625 3.500 .0065 71.5 .222 5.66 .014

Where

f’c

E c

fs

fvh

fl

f

L cc

€ max

compressive strength of concrete at time of testing,

modulus of elasticity of concrete

volumetric ratio of spiral reinforcement to the confined core

yield strength of spiral reinforcement

effective lateral confining pressure on concrete

maximum strength of confined concrete

maximum allowable concrete strain

Equations used for determining confined concrete properties are presented in

Reference 15.

* This value was obtained from Fig. C.5 using a 0.002 offset.

68
Table 4.2: Longitudinal Reinforcing Steel Details for Full-Scale and

1/6-Scale Model Bridge Column Specimens.

Specimen

f

y

(ksi) , f su

(ksi)

Bar Dia.

(in.)

number

of bars

f e sh csu Es (ksi)

Esh

(ksi

)

Models 64.7 73.2 0.276 25 .02 .0128 .090 29615 802

FullScale 68.9 103.6 1.693 25 .02 .0122 .155 27438 1454

Where

f v = yield strength of longitudinal reinforcing steel

f su

= ultimate strength of longitudinal reinforcing steel

= volumetric ratio of spiral reinforcement to the confined

G sh = stee l strain at onset of strain hardening

c su = maximum steel strain

E s = Young's modulus of reinforcing steel

E s ^

= strain hardening modulus at

core

Table 4.3: Calculated Ideal Strengths of Full-Scale and 1/6-Scale Model

Bridge Column Specimens.

?e M i Column E

i

Specimen

(kips) f ' A

cA

g

(kip-ft)

Height

( inch) (kips)

N1 26.87 . 100 32.37 29.5 13.17-

N2 53.75 .210 35.37 29.5 14.39

N4 26.87 . 100 32.37 29.5 13.17

N5 53.75 .200 35.37 29.5 14.39

SHEAR 1000.00 .071 7988.3 180.0 532.55

N3 26.87 .100 32.37 59.0 6.58

N6 26.87 .110 32.37 59.0 6.58

FLEX 1000.00 .068 8041.1 360.0 268.04

Where

P

e = applied axial column load

Mj_ = ideal moment capacity calculated using ACI approach

where an ultimate concrete strain of .003 together

with a rectangular concrete stress block are assumed.

= ideal lateral load capacity = ( M ^

- P

e

A )/h^ in general;

evaluated here at *=0 , where h^ = effective height of

test column from fixed support to point of contraf lexure

.

69
v X

I

LA

COMPRESSION

(kips)

IDEAL MOMENT CAPACITY (kip-ft)

Figure 4.1: Interaction Diagram for Full-Scale Columns

70
e309

269

228

188

147

107

66

26

0

-15

f c = 3500 psi

f y = 60,000 psi

Pbal = 92 kips

10 14 19 24 29 33

IDEAL MOMENT CAPACITY (kip-ft)

Figure 4.2: Interaction Diagram for Model Columns

71
after cycling at 2 Ay for the shear column, it was not possible to construct

moment and shear histories from the bottom up. The top-of -column reaction

force approach was therefore adopted. The equilibrium equations are

described in Appendix A (reproduced here as equations 4.1 and 4.2).

M - PafOn) . pra (h 2 ) +

P a (At ) + Wih(Ar/ 2

) (4.1)

V - P a f

- f\n < 4 - 2 )

Load histories for P a £ ( A-Frame load) , Pm (lateral load absorbed by the

vertical test machine) , and P ram (applied lateral load) for the flexure

column are shown in Figure 4.3. Note that the measured A-Frame load is

actually greater than the applied ram load due to the small offset lever

from the A-Frame attachment hinge to the test machine loading head. Load

dissipated into the vertical test machine (c) is used in conjunction with AFrame load for final determination of column base moment (d)

.

The x-axis for all of these plots is the data acquisition scan number which

serves only to separate the cyclic data for easier interpretation. P a (applied axial load) and Wj_h (weight of the column) were constant, as were

hq (height of test column from base to point of counterflexure) and h.2 (height from base of column to centerline of top-of -column hinge) . The

effective total displacement, A t , for computation of P-A moment can be

derived from Ar , the lateral displacement of column base, by consideration

of the test apparatus geometry, as shown in Appendix A, Figure A. 33.

It is interesting to note that it is possible to recover the cyclic

variation in average roller friction as follows by permutation of equations

A.l through A. 6, presented in Appendix A.

^af ^ram ^af ' ^m "^ram

8 (W + P )

i=l

This variation of average coefficient of friction is plotted versus scan

number in Figure 4.4. From this it can be seen that the coefficient of

72
friction is constantly changing during lateral displacement, as would be

expected given the friction force - displacement calibration history presented

in Figure A. 13 (Appendix A). The sharp peaks for each cycle represent the

transition from rolling to static friction as the lateral actuator reaches

the end of a lateral excursion, prior to displacement reversal. Also note

the general trend for increased coefficient of friction as the displacement

is increased. This occurs because of the greater transfer of load to the

rollers at either end of the base block (one side unloads, one side loads)

at successively greater displacements until the maximum column moment begins

to degrade. The short horizontal steps on the coefficient of friction

plot, approximately halfway from zero to the peaks, represent the

coefficient of rolling friction which averaged approximately 1% during the

test. The static coefficient of friction averaged between 2-2.5%.

Table 4.4 summarizes the column moments due to both lateral load and P-A

effects. The moments are calculated end points (maximum displacement) for

each cycle until failure using equation 4.1. Plots of load and moment

versus column lateral displacement (base of column displacement relative to

point of contraflexure ) are presented in Figures 4.5 and 4.6. Significant

drops in lateral load during an outward excursion signify points where

either spiral or longitudinal reinforcing steel within the plastic hinge

region fractured.

Figures 4.7 and 4.8 compare performance of the full-scale flexure column

with its 1/6-scale models. Lateral load has been normalized by the

calculated ACI lateral load which would cause a moment of at the

column base: thus M/M^ would have been equally applicable for the y-axis

label. This convention will be used throughout the remainder of this

report. The x-axis in Figures 4.7 and 4.8 has likewise been normalized by

dividing by the yield deflection, Ay, and thus represents displacement

ductility, /i. This permits direct comparison between model and prototype

without regard to scaling of results. The maximum moments for both full

scale column and model columns constructed from microconcrete and ready mix

concrete (N3 , N6) respectively were determined to be 1.20, 1.23, and 1.14

times the calculated ultimate moment using ACI procedures. Ultimate

73
Table 4.4 Moments for Full Scale Flexure Column

Cycle^

Lateral

Load

(kips)

Delta”

( in. )

Axial

Load

(kips)

Moment

due to

P-Delta

(k-ft)

Moment

due to

Lateral

Load

(k-ft)

Total

Moment

(k-ft)

11 E 176.30 2.743 1000.00 228.58 5289.00 5517.58

11 W 194.60 3.071 1000.00 255.92 5838.00 6093.92

21 E 268.40 7.745 1000.00 645.42 8052.00 8697.42

21 W 282.00 7.717 1000.00 643.08 8460.00 9103.08

22 E 268.20 7.772 1000.00 647.67 8046.00 8693.67

22 W 276.60 7.706 1000.00 642.17 8298.00 8940.17

31 E 282.10 11.330 1000.00 944.17 8463.00 9407.17

31 W 270.00 11.570 1000.00 964.17 8100.00 9064.17

32 E 271.70 11.630 1000.00 969.17 8151.00 9120.17

32 W 259.90 11.570 1000.00 964.17 7797.00 8761.17

41 E 269.60 15.520 1000.00 1293.33 8088.00 9381.33

41 W 255.70 15.410 1000.00 1284.17 7671.00 8955.17

42 E 260.50 15.480 1000.00 1290.00 7815,00 9105.00

42 W 248.80 15.460 1000.00 1288.33 . 7464.00 8752.33

43 E 256.90 15.480 1000.00 1290.00 7707.00 8997.00

43 W 247.70 15.450 1000.00 1287.50 7431.00 8718.50

44 E 254.30 15.470 1000.00 1289.17 7629.00 8918.17

44 W 247.50 15.440 1000.00 1286.67 7425.00 8711.67

45 E 253.00 15.470 1000.00 1289.17 7590.00 8879.17

45 W 244.80 15.440 1000.00 1286.67 7344.00 8630.67

46 E 250.80 15.470 1000.00 1289.17 7524.00 8813.17

46 W 244.20 15.440 1000.00 1286.67 7326.00 8612.67

47 E 249.50 15.470 1000.00 1289.17 7485.00 8774.17

47 W 243.50 15.450 1000.00 1287.50 7305.00 8592.50

48 E 248.00 15.470 1000.00 1289.17 7440.00 8729.17

48 W 242.50 15.440 1000.00 1286.67 7275.00 8561.67

74
Table 4.4 (continued)

Moment Moment

Lateral Axial due to due to Total

Cycle Load Delta Load P-Delta Lateral Moment

(kips) ( in. ) (kips) (k-ft)

Load

(k-ft) (k-ft)

49 E 247.70 15.470 1000.00 1289.17 7431.00 8720.17

49 W 240.00 15.450 1000.00 1287.50 7200.00 8487.50

410 E 247.30 15.470 1000.00 1289.17 7419.00 8708.17

410 W 238.40 15.450 1000.00 1287.50 7152.00 8439.50

51 E 267.60 19.380 1000.00 1615.00 8028.00 9643.00

51 W 246.00 19.380 1000.00 1615.00 7380.00 8995.00

52 E 262.30 19.390 1000.00 1615.83 7869.00 9484.83

52 W 237.90 19.370 1000.00 1614.17 7137.00 8751.17

53 E 261.90 19.440 1000.00 1620.00 7857.00 9477.00

53 W 226.60 19.410 1000.00 1617.50 6798.00 8415 . 50

61 E 242.10 23.370 1000.00 1947.50 7263.00 9210.50

61 W 223.00 23.250 1000.00 1937.50 6690.00 8627.50

62 E 242.10 23.370 1000.00 1947.50 7263.00 9210.50

62 W 197.30 23.330 1000.00 1944.17 5919.00 7863.17

63 E 186.60 23.380 1000.00 1948.33 5598.00 7546.33

63 W 120.00 23.280 1000.00 1940.00 3600.00 5540.00

6.6 E 108.10 25.630 1000.00 2135.83 3243.00 5378.83

6.6 W 20.27 25.390 1000.00 2115.83 608.10 2723.93

1 11 E = 1 Cycle 1 east excursion, 102 W = 10 *y, Cycle 2, West

excursion/ etc

.

2 This is the displacement of the centerline of the top-of -column

hinge (see Figure 2.36) at maximum extension for a given cycle.

This is used in the determination of the P-* portion of the total

moment. The actual column lateral deflection, from the base

to the point of contraf lexure , may be obtained by multiplying

the above deflections by 0.9114.

75
RAM LOAD FOR FLEXURE COLUMN TEST

A-FRAME LOAD FOR FLEXURE COLUMN TEST

Figure 4.3: Reaction Forces at work During the Testing of the FullScale Flexure Column Specimen.

76
MOMENT

(K-ft)

LOAD ABSORBED BY TEST MACHINE

FOR FLEXURE COLUMN TEST

MOMENT FOR FLEXURE COLUMN TEST

Figure 4.3: Reaction Forces at work During the Testing of the FullScale Flexure Column Specimen (continued)

.

77
COEFFICIENT

OF

FRICTION

(NU)

FRICTION COEFFICIENT FOR FLEXURE COLUMN TEST

Figure 4.4: Variation of Coefficient of Friction for Supporting Roller

System, Full Scale Flexure Column Test. X-Axis Represents

Data Acquisition System Scan Number. Generally there

were approximately equal number of scans per cycle so

that individual complete cycles are represented by two

sign reversals on the y-axis.

78
displacement ductilities of 6.6, 5, and 4 were recorded for the full-scale

column, and models N3 and N6 respectively. Note that maximum normalized

displacements were approximately 5 and 7 for models N3 and N6 (Figure 4.8).

However, these latter values were recorded after moment resistance had

dropped below 0.8 t^, the criteria used in this report as the gage for

determining when a column was ''failed” . Although currently there are no

requirements in the United States for column ductility (only confining

reinforcement ratios are required) , at least one country (New Zealand) has

adopted design spectra based on ductility with a ductility factor of 6

generally being considered the minimum value. The full-scale flexure column

therefore exceeded this criterion but the models did not.

79
LOAD

DEFLECTION

PLOT

FOR

FLEXURE

COLUMN

TEST

C\lC

CD ^ CO

. o Q- — ^ X

c\j CO c

0-

rC LO

CM CM

CO CO

cm in

CD

CD

O

o

II II

o

in

coo

o

o

in tCD CO < < 5 5 I §

& m

E

E

id

CM

ii c

LU

LU

O

<

_l

CLcnQ

(sdp« avoi ivy^ivi

80

Figure

4

. 5

Lateral

Load

vs.

Displacement

plot

for

Full-Scale

Flexure

Column.

Cycle

Numbers

Compare

with

those

described

in

Chapter

3.
MOMENT

(K-ft)

MOMENT DEFLECTION PLOT FOR FLEXURE COLUMN TEST

10,000

6000

2000

0

-2000

-6000

- 10,000

- 24.0 - 14.4 - 4.8 0 4.8 14.4 24.0

DISPLACEMENT (in)

Figure 4.6: Moment versus Lateral Displacement plot for

Full-Scale Flexure Column Test.

81
v/v,

1.5

0.9 -

0.3 -

- 0.3 -

0.9 -

1.5

4 -2 0 2 4

DISPLACEMENT DUCTILITY

Figure 4.7: Normalized Lateral Load (V/V^) versus Displacement

Ductility (A/Ay) for Full-Scale Flexure Column.

82
-5 -4 -3 -2 -1 0 1 2 3 4 5

DISPLACEMENT DUCTILITY

Figure 4.8: Normalized Lateral Load (V/V i ) versus Displacement

Ductility (A/Ay) for Model Columns N3 and N6

.

83
4.2 Shear Column Moments and Displacements

Similar to the approach uesd in Section 4.1, the information for determining

moment and shear for the full-scale shear column are shown in Figure 4.9.

The variation of average coefficient of friction is plotted versus scan

number in Figure 4.10. Both Figures 4.9 and 4.10 are similar to those

presented for the flexure column with the exception that the load

transferred to the vertical testing machine was substantially less (Figure

4.9c) than that for the flexure column and the lateral load is slightly more

than double due to the 50% reduction in column height. The reason for the

difference in the load transferred to the test machine is twofold: yield

displacement (3.53 in. for the flexure column versus 1.40 in. for the shear

column [89 mm, 35 ram]) was substantially lower (and hence hinge displacement

and therefore test machine loading head displacement) was less; and, special

measures were taken to eliminate friction between the hinge and test machine

crosshead. Knowing that the test machine would be stiffer at lower

elevations, a layer of 600 - 1.5 in (38 mm) ball bearings were placed

between two hardened steel, machined plates in molybdenum disulphide grease

atop the column. The effect is readily apparent in Figure 4.9c where it is

seen that the load absorbed by the test machine is reduced to zero over the

majority of the cycles. The spikes represent the transition from rolling to

static friction at maximum lateral displacement for each cycle. This

decrease in load absorbed by the testing machine also permits a better

picture of friction forces in the supporting roller system (Figure 4.10)

where the rolling friction portion of a cycle is now clearly evident as a

mid-height plateau at approximately +/- 1% friction. The maximum average

static coefficient of friction was approximately 2%.

Table 4.5 summarizes the column moments, due to both lateral load and P-A

effects at the maximum displacement end points of each cycle until failure

for the full-scale shear column. Moments are calculated using equation 4.1.

84
Table 4.5 Moment for Full Scale Shear Column

Cycle

Lateral

Load

(kip)

Delta^

( in.

)

Axial

Load

(kip)

Moment

due to

P-Delta

(k-ft)

Moment

due to

Lateral

Load

(k-ft)

Total

Moment

(k-ft)

11 W 407.80 1.367 1000.00 113.92 6117.00 6230.92

11 E 381.40 1.136 1000.00 94.67 5721.00 5815.67

21 W 595.80 3.342 1000.00 278.50 8937.00 9215.50

21 E 634.50 3.345 1000.00 278.75 9517.50 9796.25

22 W 563.10 3.358 1000.00 279.83 8446.50 8726.33

22 E 622.10 3.359 1000.00 279.92 9331.50 9611.42

41 W 618.80 6.777 1000.00 564.75 9282.00 9846 . 7 5

41 E 738.00 6.824 1000.00 568.67 11070.00 11638.67

42 W 611.40 6.785 1000.00 565.42 9171.00 9736.42

42 E 700.50 6.715 1000.00 559.58 10507.50 11067.08

61 W 654.10 10.220 1000.00 851.67 9811.50 10663.17

61 E 703.40 10.030 1000.00 835.83 10551.00 11386.83

62 W 566 . 70 10.050 1000.00 837.50 8500.50 9338.00

62 E 662.60 10.040 1000.00 836.67 9939.00 10775.67

81 W 607.30 13.370 1000.00 1114.17 9109.50 10223.67

81 E 674.50 13.380 1000.00 1115.00 10117.50 11232.50

82 W 565.60 13.390 1000.00 1115.83 8484.00 9599.83

82 E 653.70 13.380 1000.00 1115.00 9805.50 10920.50

101 W 574.90 16.720 1000.00 1393.33 8623.50 10016.83

101 E 623.30 16.730 1000.00 1394.17 9349.50 10743.67

102 W 544.60 16.700 1000.00 1391.67 8169.00 9560.67

102 E 498.90 16.740 1000.00 1395.00 7483.50 8878 . 50

103 W 359.60 16.680 1000.00 1390.00 5394.00 6784.00

103 E 259.00 16.740 1000.00 1395.00 3885.00 5280.00

1 11 E = 1 *y. Cycle 1 east excursion, 103 W = 10 Ay, Cycle 3 west

excursion, etc.

2 This is the displacement of the centerline of the top-of -column

hinge (see Figure 2.36) at maximum extension for a given cycle.

This is used in the determination of the P-a portion of the total

moment. The actual column lateral deflection, from the base

to the point of contraf lexure, may be obtained by multiplying

the above deflections by 0.9114.

85
RAM LOAD FOR SHEAR COLUMN TEST

A-FRAME LOAD FOR SHEAR COLUMN TEST

Figure 4.9: Reaction Forces at work During the Testing of the FullScale Shear Column.

86
MOMENT

(K-ft)

LOAD ABSORBED BY TEST MACHINE

FOR SHEAR COLUMN TEST

MOMENT FOR SHEAR COLUMN TEST

Figure 4.9: Reaction Forces at work During the Testing of the FullScale Shear Column Specimen (continued)

.

87
COEFFICIENT OF FRICTION FOR SHEAR COLUMN TEST

Figure 4.10: Variation of Coefficient of Friction for Supporting Roller

System, Full Scale Flexure Column Test. X-Axis Represents

Data Acquisition System Scan Number. Generally there

were approximately equal number of scans per cycle so

that individual complete cycles are represented by two

sign reversals on the y-axis.

88
Plots of load and moment versus column lateral displacement (base of column

displacement relative to point of contraflexure ) are presented in Figures

4.11 and 4.12. Significant drops in lateral load during an outward

excursion signify points where either spiral or longitudinal reinforcing

steel within the plastic hinge region fractured.

Figures 4.13, 4.14 and 4.15 compare performance of the full-scale shear

column with its 1/6-scale models, including those for which higher axial

loads were applied (0.2f' cAg versus 0.1f' cAg). Lateral load has been

normalized by the calculated ACI lateral load which would cause a moment

of at the column base, thus M/Mi would have been equally applicable for

the y-axis label. The x-axes in Figures 4.13 through 4.15 have likewise

been normalized by dividing by yield deflection, Ay, and thus represents

displacement ductility /i . The maximum moments for both full scale column

and the model columns (Nl, N4) which were constructed from microconcrete and

ready mix concrete, respectively, were determined to be 1.18, 1.16, and 1.20

times the calculated ultimate moment using ACI procedures. Models N2 and N5

(identical to Nl and N4 , respectively, but with the higher axial load

ratio), showed peak M/M^ ratios of 1.32 each, substantially higher than for

the lower stressed columns. The moment enhancement for those columns with

Pg/f'c^g =* 0.1 approximately agrees with the value of 1.13 predicted in

Reference 17 for columns with P e /f' cAg < 0.1. The 1.32 values for

specimens N2 and N5 are on the high side of the 15% error band predicted in

Reference 17 as follows:

M/Mi = 1 - 13 + 2.35 ( pe/ f 'cA

g

• °- 1 ) 2 ~ 1 - 15 (4.4)

Ultimate displacement ductilities of 10, 8, and 10 were recorded for the

full-scale column, and models Nl and N4 respectively. Note that maximum

normalized displacements were approximately 10 and 12 for models Nl and N4

(Figure 4.8). However, these latter values were recorded after moment

resistance had dropped below 0.8 M^; thus they were not considered in the

evaluation. All of the shear column specimens, both model and prototype,

had ultimate displacement ductilities in excess of 6, and thus would appear

89
LOAD

DEFLECTION

PLOT

FOR

SHEAR

COLUMN

TEST

CM.c

CO

O'

CM

CD <

CMc o Er~ in

CM

05 CM E

05

T—

in

CD

Tf O’ CD o o in

in o d 5 CM

ii ii ii II ii ii w< a Ps 79

u.

C/5 i_

03 c

'% CD

'r_

in

O’

N.CO

05

c\i

05

c\i

CO

LO

o°

LU

LU

o

<

_1

Qcn

o

in

o o o o O o v n n n n n 1 h- O’ f— O’ r"-

(Sdw) QV01 1VH31V1

90
MOMENT

(K

-ft)

MOMENT DEFLECTION PLOT FOR SHEAR COLUMN TEST

12,000

7200

2400

0

-2400

-7200

- 12,000

14.5 - 8.7 - 2.9 0 2.9

DISPLACEMENT (in)

8.7 14.5

Figure 4.12: Moment versus Lateral Displacement plot for

Full-Scale Shear Column Test.

91
1

A/A

2.0

-12 -8 -4 0 4 8 12

DISPLACEMENT DUCTILITY

Figure 4.13: Normalized Lateral Load (V/V^) versus Displacement

Ductility (A/Ay) for Full-Scale Shear Column.

92
V/Vj

V/Vj

1.5

-12 -8 -4 0 4 8 12

DISPLACEMENT DUCTILITY

DISPLACEMENT DUCTILITY

Figure 4.14: Normalized Lateral Load (V/V^) versus Displacement

Ductility (A/Ay) for 1/6 Scale Model Shear Columns

~ ‘ .) « 0 1 cA

g93
DISPLACEMENT DUCTILITY

Figure 4.15: Normalized Lateral Load (V/V^) versus Displacement

Ductility (A/Ay

) for 1/6 Scale Model Shear Columns

P e /(f' cA

g

) " ~ 0.2

94
to be sufficiently reinforced. The amount of spiral reinforcement (P s

)

provided for these columns, however, was substantially in excess of thatrequired by Reference 6 with the final choice for p s for the shear column

specimens being based on recommendations from CALTRANS [4]

.

Model specimens N2 and N5 (which had a P e /f c^g va^- ue 0.2) had ultimate

displacement ductilities of 10 and 12, respectively, which can be compared

with values of 8 and 10 for specimens N1 and N4 (P/f' cA

g = 0.1). This

indicates that, within the bounds studied in this test program, increased

axial load appears to be beneficial. This hypothesis is further supported by

comparing the load- displacement histories in Figures 4.14 and 4.15 which show

consistently wider hysteresis loops for a given displacement ductility,

indicating that more strain energy per cycle is dissipated inelas tically . The

ultimate moment capacities and displacement ductilities described above are

summarized in Table 4.6.

Table 4.6 Experimental Maximum Moment and Ductility Capacity

Specimen pe

f cA

g

Aspect

Ratio

(hpD)

Ultimate Moment

M- M

, i . /exp

(k-ft. ) (k-ft.

)

Capacity

M-ll /M i

; exp

Ductility Capacity

A

y

*u y=VA

y

(in.) (in.)

N1 0.100 3 32.37 38.35 1.18 0.38 3.04 8.0

N2 0.210 3 35.37 46.52 1.32 0.22 2.20 10.0

N3 0.100 6 32.37 39.76 1.23 1.01 4.04 4.0

N4 0.100 3 32.37 37.48 1.16 0.21 2.10 10.0

N5 0.200 3 35.37 46.61 1.32 0.19 2.28 12.0

N6 0.110 6 32.37 36.87 1.14 0.66 3.30 5.0

Shear 0.071 3 7988.30 11640.00 1.46 1.40 14.02 10.0

Flexure 0.068 6 8041.10 9643.00 1.20 3.53 23.30 6.6
4.3 Comparison of Test Results with New Zealand Design Charts for Structural

Ductility

It is useful at this point to make a comparison between the column ductilities

measured in this study and their predicted ductility. Predicted ductilities are

determined graphically from a design diagram presented in Reference 17, which is

based on the New Zealand code, reproduced here as Figure 4.16. The x-axis is

the ratio of the volumetric ratio of confining reinforcement provided over that

specified in NZ-3101 [20]. This ratio is represented by the letter a. The a

values for the tests reported in this paper are shown in Table 4.7. Also shown

in the table are the observed experimental ultimate displacement ductility and

the ductility predicted using Figure 4.16. The dark circles in Figure 4.16

represent the actual ultimate displacement ductilities observed during tests.

The lines in Figure 4.16 show how the predicted values were obtained for the

same specimens. The upper line is for columns with aspect ratios of 3 (the

shear columns) the lower line for an aspect ratio of 6 (flexure columns ). It

is interesting to note that the ductility of the full-scale flexure specimen,

which had approximately the amount of confining reinforcement required by NZ3101, agrees well with the predicted ductility value. The predicted ductilities

for the flexure model specimens were, however, substantially greater than those

observed experimentally. The same was true for both the prototype and model

shear specimens. One possible cause for the discrepancy, in the more heavily

confined shear specimens could be that large increases in the amount of

confining steel become increasingly ineffective and asymptotically approach some

maximum ductility. Experimental evidence to support this theory is contained

in Reference 19 which shows that vastly increased amounts of spiral confinement

in post- tensioned anchorage zones do not proportionally improve their ultimate

load capacity.

96
Figure 4,16 Structure Ductility Capacity Related to Confinement Ratio

97
Table 4.7 Effectiveness of Confining Reinforcement: A Comparison of CALTRANS

and NZ3101 Required Spiral Reinforcement and Experimental Ductility Factors.

Specimen Pe

Ratio

(hi/D) fprv fcode

a=

fprv/ fcode 4exp 4nz 4exp/4nz f ' cA

g

N1 0.100 3 .0145 .0042 3.40 8.0 17.5 0.46

N2 0.210 3 .0145 .0050 2.86 10.0 15.5 0.65

N3 0.100 6 .0069 .0041 1.63 4.0 7.5 0.53

N4 0.100 3 .0145 .0042 3.40 10.0 14.1 0.71

N5 0.200 3 .0145 .0052 2.75 12.0 15.0 0.80

N6 0.110 6 .0069 .0038 1.76 5.0 8.0 0.63

Shear 0.071 3 .0152 .0050 2.96 10.0 16.0 0.63

Flexure 0.068 6 .0065 .0052 1.22 6.6 6.1 1.08

where

(h^/D) = column height (to point of contraf lexure ) divided by diameter

fprv = spiral volumetric reinforcement specified by CALTRANS [18]

for the present test series. Note that for specimens N1,N2,

N4 , N5 , and the full scale shear column, fprv was substantially

greater than that called for in Reference 6.

fcode = spiral volumetric reinforcement required by NZ-3101 [20] = the greater of :

= 0.45 (A

g

/Ac

- 1) (f

c /fyh ) (0.5 + 1.25Pe /(*f' cA

g

))

or

= 0.12 ( f

' c / fyh ) (0.5 + 1.25Pe/(*f' cA

g

))

4exp “ experimental ultimate ductility

4nz = Predicted ultimate ductility from Figure 4. 16

& = strength reduction factor for confined members = 0.9

98
4.4 Strain in Longitudinal Reinforcement

Figures 4.17 through 4.24 show longitudinal reinforcing strains superimposed on

an outline of the test specimen. The vertical axis represents the elevation of

a specific strain gage above or below (negative units) the column base. For

reference, the axial and lateral load directions are shown on each figure. Model

specimens were oriented such that they were displaced along a north- south

direction while the full scale specimens were displaced along an east-west

direction. These directions are noted only because they were used in Chapter 3,

as well as Reference 1, in the cycle -by-cycle descriptions of the column tests;

specimen behavior obviously is not dependent on magnetic bearing.

Each specimen contained strain gages on the four longitudinal reinforcing bars

which were furthest from the neutral axis, thus providing redundant measurements

for both sides. Following the specimen orientation convention described

above, the full scale specimens contained east and west longitudinal bar gages;

models contained north and south longitudinal bar gages.

The longitudinal bar yield strain for the models and prototypes were 2184 and

2511 microstrain, respectively. Tables 4.8 and 4.9 list the depth to which the

longitudinal reinforcement yielded in the column foundation and the height above

the column base to which yielding progressed prior to column failure.

Two trends are observed in the data in Table 4.8. First, increased aspect ratio

(h]_/D) leads to nominally decreased yield length of longitudinal reinforcement

in the foundation. Increasing column axial load, on the contrary, tends to

increase the depth of yielding. The maximum depth of yielding for all cases is

approximately half of the column diameter. The experimental yield lengths

varied between 19-38% of the ACI specified development lengths.

99
Table 4.8: Extent of Yielding of Longitudinal Reinforcement

in Column Foundation

Specimen

Yield Dis.

(in.) **

D

( in. ) YE/D id

(in)

YD/ld

N1 3.3 9.8 .34 15.3

* .21

N2 4.7 9.8 .48 15.67' .30

N3 2.8 9.8 . 2*9 14.9

'

. 19

N4 4.8 9.8 .49 15 . 27' .32

N5 5.4 9.8 .55 15.2',' .36

N6 3.8 9.8 .39 15.5'' .24

SHEAR 31.7 60.0 .53 83.0 .38

FLEXURE 25.6 60.0 .42 81.2 .31

* Scaled from prototype development length using 1^ = 0.085 f v

/Vf’ c and a scale factor of 6.1. ** Average of four values representing the peaks for the east and west

(north and south, models) instrumented bars at their displacement

maxima to either side of center position.

Where

YE = length of longitudinal reinforcement embedded in foundation

over which yielding has occurred (inches)

1 D = ACI development length (inches) for longitudinal reinforcement

D* = column diameter (inches)

100
1000 K

V

1 *7 / OC>PAF —350

1000 K — PAF

350 —

Figure 4.17: Longitudinal Reinforcement Strains for

Full-Scale Flexure Column. Top Figure

= East Gage Set; Bottom = West Gage Set

101
n x Ay Cycles

~1000 K

Figure 4.18: Longitudinal Reinforcement Strains for

Full-Scale Shear Column. Top Figure

= East Gage Set; Bottom = West Gage Set

102
Figure 4.19: Longitudinal Reinforcement Strains for

Model N1 Top Figure = North Gage Set;

Bottom = South Gage Set

103
Figure 4.20: Longitudinal Reinforcement Strains for

Model N2 Top Figure = North Gage Set;

Bottom = South Gage Set

104
I i—20

PAX

9000

Figure 4.21: Longitudinal Reinforcement Strains for

Model N3 Top Figure = North Gage Set;

Bottom = South Gage Set

105
Figure 4.22: Longitudinal Reinforcement Strains for

Model N4 Top Figure = North Gage Set;

Bottom = South Gage Se.t

106
- 5000_.

- 10,000 -7500

STRAIN (microstrain)

7500

-..--4 I -

10,000

-(--10

-)—20

_i

Figure 4.23: Longitudinal Reinforcement Strains for

Model N5 Top Figure = North Gage Set;

Bottom = South Gage Set

107
n x Ay Cycles 60<^>

Figure 4.24: Longitudinal Reinforcement Strains for

Model N6 Top Figure = North Gage Set;

Bottom = South Gage Set

9000

1 9000

108
4.5 Strains in Confining Reinforcement

Figures 4.25 through 4.32 present strain profiles for the confining

reinforcement in a manner analogous to that for the longitudinal reinforcement.

These generally indicate, that there is minimal yielding of the spiral inside

the foundation block and that most of the confining stresses are concentrated

between the base of the column and a height approximately one column diameter

above the base. Above this elevation there is a rapid decrease in confining

steel strain to well below yield levels even at the highest displacement

ductilities. No clear differences existed between specimens with different

column aspect ratios and differing axial loads.

4.6 Column Curvatures and Estimation of Plastic Hinge Lengths

If the axial strains are known at two locations on opposing sides of the neutral

axis at the same elevation above the column base, then the column curvature at

that point is given by:

1/R = (e

x

- e 2 ) /X (4.5)

where and €2 are the Cwo measured strains and X is the separation distance

between the two instrumented locations. Generally, X is taken to be the column

diameter, D, and the strains can be measured either by means of strain gages on

longitudinal reinforcement or by means of external potentiometers. The latter

technique is useful provided a stiff mounting mechanism is available that is

unaffected by concrete spalling. This technique is more resilient and can be

used to observe variations in curvature to lateral displacements near failure.

The strain gage approach was used initially and generally produced reliable

Information for only a few cycles beyond a displacement of 2Ay , at which point

the gages typically debonded at their limiting strain. The external clip gage

technique was implemented for the second full scale test and produced reliable

results up to a displacement of 6Ay. At that point the lower gages were removed

so as to prevent damage from spalling concrete.

Plots of column curvature are presented in Figures 4.33 through 4.37.

109
1000 K \7

P AF

I -j— 350

300

<^>

1000 K

~i u

<CrrD>pAF —350

—300

n x Ay Cycle —250

— 200

Figure 4.25: Confining Steel Strain for Full-Scale

Flexure Column. Top Figure = North Gages

Bottom Figure = South Gages.

110
n x Ay Cycles

1000 K

-6000 /

-4400 -2800 -1200

\ STRAIN (microstrain) ^

— 200

1

1

-t-150

l

1

1

1

1 1-100

1

1

!

1

1

1

-f 50

l

' f /

i )\( Vi

1 A 1 ‘*V .*V

PAF

<5=0

2800

W

4400

6000

-—50

-i—100

A.

n x Ay Cycles

124

10

n 1000 K

— 200 1

— 150

— 100

Paf

<0=0

=.0

-6000 -4000

STRAIN (microstrain)

-2000

W

2000 - - +- - 4000 6000

- -50 pL

| <J=C>

--100

Figure 4.26: Confining Steel Strain for Full-Scale

Shear Column. Top Figure = North Gages;

Bottom Figure = South Gages.

Ill
+-20

Figure 4.27: Confining Steel Strain for Model

Column N1 . Top Figure = East Gages;

Bottom Figure — West Gages.

112
-1—20

Figure 4.28: Confining Steel Strain for Model

Column N2 . Top Figure = East Gages;

Bottom Figure = West Gages.

113

AlUlfl
Figure 4.29: Confining Steel Strain for Model

Column N3

114
Figure 4.30: Confining Steel Strain for Model

Column N4 . Top Figure = East Gages;

Bottom Figure = West Gages.

115
Figure 4.31: Confining Steel Strain for Model

Column N5. Top Figure — East Gages;

Bottom Figure = West Gages.

116
Figure 4.32: Confining Steel Strain for Model

Column N6 . Top Figure = East Gages;

Bottom Figure = West Gages.

117
Figure 4.33: Column Curvature (radians/inch) for

Full-Scale Flexure Column

118
I 1

Figure 4.34: Column Curvature (radians/inch) for

Model Specimens N3 and N6 , Companion

Specimens for Full-Scale Flexure Column

119
Figure 4.35: Column Curvature (radians/inch) for

Full-Scale Shear Column

120
Figure 4.36: Column Curvature (radians/inch) for

Model Specimens N1 and N4 , Companion

Specimens for Full-Scale Shear Column

121
Figure 4.37: Column Curvature (radians/inch) for

Model Specimens N2 and N5 , Companion

Specimens for Models N1 and N4 with

P e /f' cA

g

=0.2.

122
If a linear strain distribution is assumed up to first yielding of the

longitudinal reinforcing, then the yield curvature is calculated by moment area

principles as:- (3A

y

/L2

)

' (4.6)

Where Ay is the experimentally determined yield deflection, and L is the height

of the column from the base to the point of contraflexure . These values are

tabulated in Table 4,9 along with other information concerning plastic hinge

lengths. The experimental values of plastic hinge length for the tests

discussed in this report were determined as that length of column over which the

curvature exceeded 0y (see delimiting lines on Figures 4.33 through 4.37). With

the exception of the full-scale flexure column the plastic hinge length Lp was

generally between 0 . 7D to 1.2D. These values were substantially larger than

the recommended theoretical value of 0.5D given in references 15,16, and 17.

The experimentally determined value of Lp for the full scale flexure specimen

(Ip/D = 1.98) seems well outside that predicted using equations presented in

references 15, 21, and 22. The test data for this column was closely

scrutinized and appears valid, despite the discrepancy.

Note that in Table 4.9 the yield curvatures (<£y) for the model specimens have

been divided by the scale factor (S]_) of 6.1, since curvature has dimensions of

1/S]_. This was done for comparison purposes only. The delimiting lines used

to determine the bounds of the plastic hinge for the model specimens utilized

the unsealed, experimentally determined value of <f>y.

123
Table 4.9: Experimental Plastic Hinge Lengths (Lp)

Specimen cp v

3 Spall Height

y (in.)

xlO' 4

( rad . / in . )

Extent of

Yielding of

Long. Bar

( in. )

Exp

T 2 L

P

(in.)

Exp

Lp/D

Pred

Baker

( in. )

Pred

Corley

( in. )

Shear Specimens: P

e /f' cAg = 0.1

Full Scale 1.296 66.0 134.0 43.2 .72 18.6 32.5

Model N1 2.148 3 3.0 11.0 A _l„ A /\ >\ A 5.9 6,8

Model N4 1.187 2.5 8.0 7.5 .76 5.9 6.8

Shear Specimens: P

0

/

1 cAa = 0.2

Model N2 1.243 4.0 11.0 11.6 1.18 6.3 6.8

Model N5 1.074 3.0 8.0 9.6 .98 6.3 6.8

Flexure Specimens: P

0 /f ! cAg 0.1

Full Scale 0.817 50.0 119.

4 119.0 1.98 37.1 37.4

Model N3 1.427 5.0 11.0 10.2 1.05 11.7 8.7

Model N6 0.933 5.0 11.0 10.6 1.08 11.7 8,7

Notes

1 . q)y = (3*y/L^). The curvature values for the models were divided by s-j_ (scale factor) to account for scale.

2. Plastic hinge length was obtained from Figures 4.33 through 4.37 and was

determined as that length over which the curvature (as measured

experimentally) was greater than cpy.

3. Excessive flexibility for this column (* was approximately twice the

predicted value), as well as premature failure of the instrumentation used

to determine curvature, precluded an experimental determination of Lp.

4. No instrumentation was available above this height and it is possible

that longitudinal bar yielding could have extended further.

124
4.7 Inelastic Strain Energy Capacity

An effective method for measuring overall column performance subjected to cyclic

load is to calculate the amount of strain energy dissipated through inelastic

action. The two cyclic load test data analysis computer programs described in

Reference 1 (TURBO-LOOP and GRAPH) were used to determine the amount of strain

energy dissipated during each load-displacement cycle for the full scale tests.

This was done using computer graphics by filling the area bounded by the loaddeflection hysteresis loops with pixels of a specified color, counting the

number of pixels of that color, and converting this to engineering units by

means of a scale factor. Figure 4.38 shows two bar- graphs generated using this

technique for the prototype flexure and shear column tests. Several things are

apparent from these plots. First, each increase in displacement ductility

(until failure of the column) results in a substantially higher amount of energy

being inelastically dissipated. The second cycle at a given displacement

ductility generally results in the dissipation of slightly less energy. It is

interesting to note, however, that the results for the flexure specimen, which

sustained 10 cycles at 4Ay, indicate that after the third cycle at 4Ay that the

energy dissipation appears to stabalize so that the 10th cycle dissipates nearly

90% that of the first. This is important because it indicates that there

apparently exists a threshold displacement below which a substantial number of

inelastic cycles may take place during which energy is dissipated without marked

degradation of the column. Above this threshold, additional cyclic load quickly

causes failure. This behavior is evident in both the flexure and shear

columns by the rapid decrease in energy per cycle during the final few cycles

before the column failed.

4.8 Similitude Relationships: Model versus Prototype

Before discussing specific findings regarding similitude it will be useful to

briefly review the fundamental concepts of scale factors for reinforced concrete

structures. A "true model" is one which maintains complete similarity with

the prototype. Stress and strain characteristics for the materials used in a true model are scaled by the factors s a and s

£

, respectively, and lengths are

modified by the previously mentioned length scale factor, s^. For reinforced

125
concrete, it is not practical to achieve this due to the difficulty of scaling

the modulus of elasticity of the reinforcing bars. Steel is the only practical

reinforcing material. This leads to the concept of a "practical true model", in

which s a

=* s

£ = 1 and the consequent requirement that model reinforcing steel be

capable of reproducing the stress - strain curve for the prototype reinforcement.

Ideally, the concrete stress - strain curve should also replicate that for the

prototype. But because of different curing conditions this is generally not

possible and a "distorted model" results. However, a practical true model can

still be achieved by scaling the resultant forces and moments by s Q. A

practical experimental value of s^ is the ratio of the compressive strengths of

prototype and model concretes, f' C p/f' cm . What remains, theoretically, is the

relationship between model and prototype based solely on the length scale factor

S]_. In this report, estimates of s^ are determined experimentally from

consideration of deflections, moments, and strain energy. The following

symbols and definitions will be used throughout the remainder of the report:

s ld 3

s lm 3

s le 3

where

>Aym

»up

^um

E

E

E

E

Pc

me

3

pt

mt 3

experimental scale factor for yield deflection

Ayp/Aym

experimental scale factor for ultimate moment

(^up/^um ) 173

(4.7)

(4.8)

experimental scale factor for absorbed strain energy

(Epc/^mc)

"^ 73 (4.9)

(Ept/Emt ) 1/3 (4.10)

yield deflection for prototype column (inch)

yield deflection for model column (inch)

ultimate moment for prototype column (kip -inch)

ultimate moment for model column (kip- inch)

cyclic strain energy for prototype column (kip- inch)

[individual cycles]

cyclic strain energy for model column (kip- inch)

[individual cycles]

total cyclic strain energy for prototype columns (kip- inch)

[summation of all cycles until failure]

total cyclic strain energy for model columns (kip- inch)

[summation of all cycles until failure]

126
18,000

sO vC

o o O o O

o o o o o

o o o o o

in CM cn CD CO

(U!->1) A0H3N3

127

Figure

4.38

Cyclic

Strain

Energy

for

Prototype

Shear

and

Flexure

Specimens.

Each

bar

represents

the

area

bounded

by

the

load-deformation

hysteresis

loop

for

successive

cycles

through

failure.
COMPARISONS OF HYSTERESIS CURVES

FOR THE FLEXURE SPECIMENS

Figure 4.39: A Comparison of Model and Prototype LoadDeformation Behavior Under Reversed Cyclic Loading

for Flexure Specimens. Note that Model Data has been

scaled by s^.

128
COMPARISONS OF HYSTERESIS CURVES

FOR THE SHEAR SPECIMENS

Figure 4.40: A Comparison of Model and Prototype LoadDeformation Behavior Under Reversed Cyclic Loading

for Shear Specimens. Note that Model Data has been O scaled by s ^_ ° .

129
The proximity of these experimentally determined scale factors to that for the

practical true model (sq — 6.1) will serve as a measure of the ability to

predict prototype bridge column performance from model tests.

The units of strain energy (both cyclic and total, as defined above) are force

multiplied by displacement. To achieve similitude force must be multiplied by

the square of si (37.21) and displacement must be multiplied by si (6.1). Thus,

energy is scaled by sq J for the practical true model. To account for

differences in concrete compressive strength the model energy was also

multiplied by s a , as defined above. Figures 4.39 and 4.40 show comparisons

between load- deflection behavior for model and prototype columns for each cycle.

When viewing these figures keep in mind that model columns N1 and N3 were

constructed from microconcrete in accordance with accepted experimental practice

[23]. Models N4 and N6 were constructed from commercially available ready-mix

concrete which used "pea gravel" (1/4 inch [6 mm] nominal maximum size washed

river gravel) as coarse aggregate. The trends shown in Figures 4.39 and 4.40

indicate that the models constructed from ready-mix concrete reproduced the

performance of the prototype quite well on a cycle by cycle basis. The

microconcrete models were generally shown to be more flexible. Thus they had

larger initial displacements when Ay was determined. This increased

displacement under similar applied lateral load explains the differences between

the shape of each hysteresis loop for the microconcrete models versus those for

the prototype and ready-mix model. In general, more energy was dissipated per

cycle in the microconcrete models, and was more pronounced in the shear column

tests

The scaled areas bounded by the hysteresis loops of Figures 4.39 and 4.40 can be

used to determine energy dissipation per cycle, using the procedures defined in

Section 4.6. These values are shown on the bar charts in Figure 4.41. These

charts clearly show that the microconcrete models overpredict the energy

dissipation per cycle in the prototype, and exhibit premature column failure, as

evidenced by the rapid degradation in energy dissipation beginning at 4 Ay for

the flexure columns and 8 Ay for the shear columns. The rapid failure of the

microconcrete flexure column was caused by fracture of the confining spiral

reinforcement in the plastic hinge region during the second

130
ENERGY

(kip

in)

.

ENERGY

(kip

in)

COMPARISON OF ENERGY ABSORBED ON A PER CYCLE BASIS . FOR THE FLEXURE COLUMN

16,000 r

14,000

12,000 -

10,000 -

8000

6000

4000

2000

rJ

|

Model N3

[] Model N6

[] Full-Scale

1

1 Ay 2Ay 3Ay 4 Ay 5 Ay 6 Ay

COMPARISON OF THE ENERGY ABSORBED ON A PER CYCLE BASIS

FOR THE SHEAR COLUMNS

27,000

24,000 -

21,000 -

18,000 -

15,000 -

12,000 -

9000 -

6000

3000 -

|

Model N1 ^

Model N4

Full-Scale

1 Ay 2Ay 4Ay 6Ay 8 Ay 10 Ay

Figure 4.41: A Comparison of Cyclic Energy Dissipation

between Model and Prototype Specimens. Top

Chart is for Flexure Specimens (h]_/D = 6)

Bottom Chart is for Shear Specimens (h]_/D = 3)

131
cycle at 4 Ay. The ready-mix model columns, on the other hand, appear to

track the cycle-by-cycle performance of the prototypes quite well, although the

ready-mix model flexure column shows some overprediction of energy dissipation

per cycle.

An important factor not accounted for in these graphs is the effect which losses

in applied lateral load (due to friction in the roller transport system, and

loads transferred to the vertical test machine) had in the determination of Ay

for the prototype columns. Following completion of the prototype tests a

comparison was made between the effective lateral load applied to the prototype

column (actual column base moment divided by column height after friction losses

had been accounted) versus the applied lateral load. The variation of fq e =

Fe ff/Fram ( w^ere Fe ff = the actual lateral load applied to the column, F ram = the lateral load applied to the column base block by the hydraulic actuator, and

f]_ e is the load effectiveness ratio) is plotted in Figure 4.42 for a typical

cycle of lateral load. The spikes represent locations where the ram load

underwent a sign reversal while a finite value of effective lateral load

remained on the column. Such load reversal spikes occurred twice per cycle.

Between the spikes a relative steady state exists over which the ratio of

effective lateral load to ram load is constant. For the prototype flexure

column this value was 0.913; for the shear column it was 0.973. If it is

assumed that both effective lateral load, and therefore the initial estimate of

Ay, were affected by these loss ratios, then the effective energy per cycle that

would have been obtained if losses were not present can be obtained by dividing

prototype cyclic energy by (f]_ e )^ This term is defined as the energy

modification factor (fem ) and is used to compensate for the fact that no losses

existed in the model tests. For the flexure column the energy modification

factor was 0.761; for the shear column it was 0.922. Revised bar graphs which

incorporate the above adjustment factor are shown in Figure 4.43. These

indicate further improvement in the relation between ready-mix models and the

prototypes. However, the microconcrete models still differ substantially from

the prototypes. One possible explanation for the difference is the importance of

aggregate interlock in cyclic energy dissipation. It is likely that less

interaction of this nature takes place in microconcrete.

132
LATERAL TO RAM LOAD RATIO FOR THE FLEXURE COLUMN

LATERAL TO RAM LOAD RATIO FOR SHEAR COLUMN,

CYCLE 11

Figure 4.42: Ratio of Effective Lateral Load to Applied

Ram Load for prototype Flexure and Shear

Columns

.

133
COMPARISON OF ENERGY ABSORBED ON A PER CYCLE BASIS

FOR THE FLEXURE COLUMN (Modified)

COMPARISON OF THE ENERGY ABSORBED ON A PER CYCLE BASIS

FOR THE SHEAR COLUMNS (Modified)

Figure 4.43: A Comparison of Cyclic Energy Dissipation between Model and

Prototype Specimens with modifications to account for

differences in compressive strength between models and

prototypes, and for friction losses in the prototype tests.

Top Chart is for Flexure Specimens (hl/D = 6);

Bottom Chart is for Shear Specimens (hl/D = 3)

134
Table 4.10 Experimental Scale Factor for Absorbed Energy"

[on a cycle-by-cycle basis]

N3 = Microconcrete flexure model, N6 = Ready-mix flexure model

2: Failure of model.

3: Failure of full scale column.

4: Averaged up to failure of the model.

5: Cycle 41, e.g., is Cycle number 1, displacement ductility of 4*y etc.

** Energy calculated by graphically integrating area bounded by lateral load

vs column deflection hysteresis loops for each complete lateral excursion at a given displacement ductility using procedures outlined in Reference 1. Cyclic

energy for the model specimens was multiplied by s Q as described in Section 4.7.

Prototype cyclic energy was scaled by f em , as described in Section 4.7.

135
Values listed represent the Cyclic Experimental Scale Factor for Absorbed Energy

( s le)

At the beginning of this section it was stated that the model energy was

scaled by s^^ for comparison in the bar charts just presented. The proximity

between model and prototype individual cycle energy values in the bar charts

therefore represents a measure of the accuracy with which prototype behavior

is predicted by model tests. For example, the high values of cyclic strain

energy dissipation for the microconcrete models, when compared with the

prototype, indicate that the scale factor for energy absorption, based on

i these tests, is less than 6.1. The values listed in Table 4.10 examine this

aspect of similitude by presenting the experimental scale factor for energy

absorption, S]_ e , calculated on a cycle -by- cycle basis. This calculation is

possible because identical load histories were used for both models and

prototypes. Averages for S]_ e up to the failure of the model specimens are

presented at the bottom of the table. An experimental scale factor less than

S]_ (which for this set of tests was 6.1) indicates that the models

overestimate the prototype absorbed cyclic strain energy. In this sense, the

microconcrete models are seen to perform poorly, with approximately a 30%

difference between experimental and practical true model scale factors.

Therefore, use of microconcrete test results (based on the limited data

available) would likely result in unconservative expectations for prototype

energy absorption capacity.

The ready-mix shear model agrees well with the prototype on a cycle-by-cycle

basis and exhibits only a 1% difference between experimental and practical

true scale factors. The ready-mix flexure model, while generally tracking

prototype behavior (see Figure 4.39), is shown to have a 12% difference

between experimental and practical true scale factors, based on consideration

of cyclic absorbed strain energy.

In Table 4.10 comparisons were made between cyclic energy dissipation for

model and prototype columns on a cycle-by-cycle basis. An alternative

comparison may be made between the total energy (summation of cycle energies)

dissipated by model and prototype columns up to their respective failure

points. Failure was determined when the maximum moment decreased to less than

136
80% of the peak moment at 2Ay, first cycle. Column (a) in Table 4.11

represents unmodified test data, with no corrections made for differences in

concrete strength between specimens, nor accounting for friction losses and

test machine interaction in the prototype tests. Columns (b) and (c) in Table

4.11 successively add these corrections. The final values [column (c)]

indicate reasonable correlation between the total energy dissipation capacity

of the ready-mix model columns and their prototype companions. The

experimental scale factor for total energy for the ready-mix shear model

column is shown to exceed the practical true model scale factor by 8%

(unconservative) while that the ready-mix flexure model column leads to a 4%

lower (conservative) value. Both of these values are encouraging, given the

variance generally associated with testing of reinforced concrete structures,

and indicate that total energy absorption capacity can likely be predicted

through the use of appropriate model tests. No comparisons are made in Table

4.11 between microconcrete model columns and their respective prototypes due

to the inability of these models to reproduce the cyclic load- deformation

behavior of the prototypes.

Table 4.12 summarizes two other important specimen parameters relating model to

prototype performance. The experimental scale factor for yield deflection,

sq^, has been determined using equation (4.7) and is shown in column (a).

Column (b) modifies sq^ to account for losses due friction and test machine

interaction for the prototype columns. The ratio of sq/sq^' is a measure of the

effectiveness with which the model data can be used to predict prototype yield

displacements. Values of this ratio greater than 1.0 indicate conservatism

(displacements predicted by model tests would be greater than actually achieved

in the prototype), whereas values less than 1.0 are unconservative (prototype

displacements would be greater than predicted on the basis of model tests)

.

Both of the above statements assume that model results would be extrapolated to

prototype using the practical true length scale factor, sq, and stress scale

factor s a

. Use of microconcrete model columns would therefore lead to

overestimation of prototype yield deflection -- by an average of 60%. The

ready-mix shear model specimen was shown to underpredict prototype yield

displacement by 11%, while the ready-mix flexure model over-predicts yield

displacement by 4%.

137
The second part of Table 4.12 presents information concerning the experimental

scale factor for ultimate moment, s^ra , determined in accordance with equation

4.8. Similar to the calculations for strain energy, two levels, of

modifications are made to the original column moment data presented in column

(c) . Differences in concrete compressive strength are accounted for in column

(d) while losses due to friction and test machine interaction in the prototype

tests are accounted for in column (e) . The ratio of s^/s^m

' ' , shown in the far

right column, indicates that all of the models (both microconcrete and readymix) were able to accurately predict prototype ultimate moment using only the

considerations for the practical true model (moment scales by sj_ J

) , and the

stress scale factor, s a

.

The ability of the microconcrete model columns to accurately predict prototype

ultimate moment, yet not be able to model deflections nor cyclic energy

dissipation, is difficult to explain. Ultimate moment is determined chiefly

from the compressive strength of the concrete and the ultimate tensile strength

of the reinforcing steel. Because both the ready-mix and microconcrete models

had similar concrete strengths, as well as identical longitudinal reinforcement,

it should be expected that they would exhibit similar ultimate moments.

Adjustment of the predicted prototype moments by the scale factor for stress,

s a , also appears valid given the uniform agreement between the resulting

experimental and length scale factors for the practical true model.

Since the reinforcement and reinforcing pattern was the same for all model

columns, the only likely explanation for a large reduction in section stiffness

(leading to increased displacement under lateral load) would be a difference in

concrete stiffness between the ready-mix concrete and microconcrete. The mix

proportions for these two concretes, presented in Table 4.13, show that the

microconcrete has a lower volume fraction of aggregate than the ready-mix

concrete. Because the modulus of the aggregate is approximately 5 times greater

than that for cement paste, the ready-mix is inherently a stiffer concrete.

Additionally, the cement paste in microconcrete contains greater air content

because the smaller size aggregates entrap more air in mixing. These factors

can be shown through simulation to produce a decrease of approximately 20% in

modulus of elasticity for microconcrete when compared to ready-mix concrete.

138
Table 4.11 Scale Factors Relating Prototype and l/byScale Model Column

Behavior: Total Strain Energy to Failure''

Specimen

(ksi)

(a)

Energy Absorbed

Up to Failure

(k-in.

)

(b) 1 Column (a)

x s a

(k-in.

)

(c) 3

Column (b) * f em

(k-in.

)

N4 3.54 348.70 518.36 518.36

Shear 4.98 83888.24 83888.24 91914.89

N6 3.38 351.08 540.12 540.12

Flexure 5.20 105923.85 105923.85 138270.80

s le s

l /s le s le ' s ]/ s le ' s le

* ' s l /s le’

Shear/N4 6 . 22 3 0.98 5 . 44 3 1. 12 5 . 62 3 1.08

Flexure/N6 6 . 7

1

3 0.91 5.81 3

1. 05 6 . 34 3 0.96

1: The energy absorbed by the model columns has been modified to account for

differences in concrete compressive strength between model and prototype.

2: The energy absorbed by the prototype column has been modified to account for

losses in lateral load due to friction and test machine interaction when

determining for the prototype columns. Note that f 0m = 1.096 for the

shear column 1 . 3054 for flexure column.

3: This value represents the cube root of the ratio of the sum of all prototype

cyclic strain energy to the sum of all model cyclic strain energy up to

failure of the respective columns (e.g. Shear/Nl). Recall that N1 and N3

were constructed with microconcrete; N4 and N6 were constructed using fine

aggregate ready-mix. Otherwise these specimens, respectively, were identical.

*: No comparison is made for microconcrete models, since these were not

effective in reproducing prototype behavior on a cycle-by-cycie basis.

139
Table 4.12 Scale Factors Relating Full-Scale and 1/6-Scale Model Column

Behavior: Displacements and Moments

Specimen f'c

(ksi)

(a)

A

v

( in.

(b) 4

x‘l/f

( in. ) le

(c)

Ultimate

Moment

(k-in.

(d) 1 Column (c)

x s

CT ( k- in .

(e) 2

Column

x l/f en (k-in.

(d)

l

N1 3.49 0.38 0.38 38.35 54.72 54.72

N4 3.54 0.21 0.21 37.48 52.73 52.73

Shear 4.98 1.40 1.44 11640.00 11640.00 12753.95

N3 3.68 1.01 1.01 39.76 56.18 56.18

N6 3.38 0.66 0.66 36.87 56.71 56.71

Flexure 5.20 3.53 3.86 9643.00 9643.00 12584.11

s ld s ld s

l /s ld’ s lm

3 s lm'3 - . . 3 s lm s

l /s lm’

Shear/Nl 3.68 3.78 1.61 6.72 5.97 6.16 0.99

Shear/N4 6.67 6.85 0.89 6.77 6. 04 6.23 0.98

Flexure/N3 3.50 3.82 1.60 6.24 5.55 6.07 1.00

Flexure/N6 5.35 5.85 1.04 6.40 5.54 6.05 1.01

1: Ultimate moment for model columns has been modified to account for

differences in concrete compressive strength between model and prototype.

2: Ultimate moment for prototype column has been modified to account for

losses in lateral load due to friction and test machine interaction.

Note that f em = 1.096 for the shear column; 1.3054 for flexure column.

3: This value represents the cube root of the ratio of the prototype moment

to the moment in its corresponding model (e.g. Shear/Nl). Recall that N1 and

N3 were constructed with microconcrete; N4 and N6 were constructed using

fine aggregate ready-mix. Otherwise these specimens, respectively, were

identical

4: Yield displacement for prototype columns has been modified to account for

losses in lateral load during determination of *y (see note 2 above)

140
This would explain the 22% greater yield displacement observed for microconcrete

model column N2 when compared with its ready-mix counterpart, N5 . But it does

not account for the 53% and 80% increase in yield deflection as indicated by the

ratios of Ay (N3/N6) and Ay (N1/N4) , respectively.

One possible source for the remaining, and highly variable, difference in

stiffness is because microconcrete mixes are typically somewhat sticky, are

considerably less workable, and exhibit much smaller slump than for ready-mix

concrete of equal compressive strength. It is thus difficult to place, even

with vibration. Visual inspection of the microconcrete specimens following test

indicated the presence of numerous small voids varying in size from 1/8 - 1/4

in. (3-6 mm) in diameter. Unfortunately, the column specimens were sufficiently

cracked following test that useable cores -- which would have allowed for the

determination of in- situ concrete modulus -- were not obtainable.

Even if the differences in stiffness described above could be quantitatively

accounted for, the rapid failure of the microconcrete columns, as evidenced in

Figure 4.43, precludes their ability to track the prototype test columns to

failure. It is the author's opinion that this shortcoming results from the

inability of microconcrete to develop significant shear force across the cement

matrix under load reversal following initial cracking -- loads carried in readymix concrete by means of aggregate interlock.

Table 4.13: Mix Design for Model Columns

Material Microconcrete Volume

(lb/cu.yard) (cu.ft.)

Ready-Mix Volume

(lb/cu.yard) (cu.ft.)

Cement (Type I)

Fine Aggregate

Coarse Aggregate

Water

slump

699.3 3.57

2724.3 15.88

0.0 0.00

461.0 6.96

0.5 in.

605.7 3.09

1460.0 8.52

1575.0 9.19

370.4 5.45

4.0 in

.

141
-

.
5 . 0 CONCLUSIONS

1. The requirements for longitudinal and confining (spiral) reinforcing steel,

as defined in Reference 6 [CALTRANS] are sufficient to achieve a displacement

ductility factor of 6 for full-scale bridge columns with an aspect ratio

(height/diameter) of 6 and a displacement ductility factor of 10 for full

scale bridge columns with an aspect ratio of 3. These values were based on

tests of single prototype specimens.

2. Similitude studies were conducted to determine if prototype behavior could

be predicted from tests of 1/6-scale model replicates. Structural modelling

theory calls for the scaling of not only the reinforcing steel, but also the

concrete, leading to the use of "microconcrete." The use of microconcrete for

specimens subjected to cyclic load, particularly those in which shear effects

might be expected to dominate cyclic behavior, was questioned due to the

possible differences in the details of the aggregate - interlock mechanism of

microconcrete compared with prototype concrete. Two sets of model specimens

were thus constructed for the present study, one using microconcrete and one

using ready-mix concrete with small size (0.25 in.; 6 mm) river gravel. The

results of these similitude tests are as follows:

a) Model specimens constructed from ready-mix concrete were able to

consistently reproduce prototype behavior, on a cycle by cycle basis,

with scale factors for absorbed strain energy, ultimate moment, and

yield displacement falling close to those for the practical true model,

s l*

b) Model specimens constructed from microconcrete were substantially more

flexible than both the ready-mix models and prototypes, and hence

exhibited substantially greater yield deflections. Failure was

generally more rapid than for ready-mix specimens and ultimate

displacement ductility was 20% less. Because of this, model cyclic

absorbed strain energy, when modified to account for scale factor and

differences in concrete strength, is shown to be overpredict

(unconservative) the observed results from prototype tests.

143
c) Both microconcrete and ready-mix models were able to accurately

reproduce prototype ultimate moment capacity. Moment comparisons

were made by scaling model ultimate moments by sp^ and multiplying

this result by the scale factor for stress, s a , to account for

differences in concrete strength.

d) The ability of the ready-mix model columns to predict with

reasonable accuracy the strain energy dissipated during each

cycle (the scaled hysteresis loopes essentially overlapped)

as well as the total energy dissipated up to column failure

shows that correct selection of materials is important for

accurate behavior of concrete models subjected to high level

cyclic loads.

3. Failure in both model and prototype column specimens was initiated by

fracture of the confining reinforcement followed by buckling of the

longitudinal reinforcement. Low cycle fatigue generally resulted in fracture

of the longitudinal reinforcement during the next level of displacement

ductility above that during which the spiral, first fractured.

4. Yielding of the longitudinal reinforcement in the column foundation was

limited to a depth of less than 50% of the column diameter. This represents a

distance approximately 40% of the specified (ACI) development length for the

longitudinal reinforcement.

5. Experimentally determined ultimate column moments were substantially

greater than the nominal (ACI) calculated moment capacity (with = 1.0). The

prototype columns (aspect ratios of 6 and 3) exhibited experimental ultimate

moment to ACI nominal moment ratios (moment enhancement ratio) of 1.20 and

1.46 respectively. Similar ratios were observed for the model specimens.

These results are for columns with axial loads of P e /f' cAg = 0.1.

6. An increase in column axial load from P e /f' cAg = 0.1 to 0.2 resulted in a

20% increase in ultimate displacement ductility and a 12% increase in moment

enhancement ratio. These were based on comparisons between four model bridge

columns and support the findings of other researchers [17].

144
7. The second cycle at a given displacement ductility generally resulted in

the dissipation of slightly less energy. It is interesting to note, however,

that the results for the flexure specimen, which sustained 10 cycles at 4 Ay,

indicated that after the third cycle at 4 Ay that the energy dissipation

appeared to stabalize so that the 10th cycle dissipated nearly 90% that of the

first. This is important because it indicates that there apparently exists a threshold displacement below which a substantial number of inelastic cycles

may take place during which energy is dissipated without marked degradation of

the column. Above this threshold, additional cyclic load quickly caused

failure

.

8. Important lessons were learned in carrying out the prototype tests,

especially with regard to the design and implementation of instrumentation for

monitoring reactions in statically indeterminate test apparatus. The most

reliable load measurement systems consisted of single load resisting elements

with redundant strain gage instrumentation for verification. Where

appropriate, the test fixtures themselves (including the vertical test machine

and tie-back frame) were instrumented, calibrated, and successfully used

during the full-scale tests. Additional ins trumentation was used to provide

two independent means of determining column moments and shear. This avoided

the possible loss of test data should one set of instrumentation become

inoperable during testing. A suggested improved loading apparatus which

makes use of single load path instrumentation is presented in Appendix B.

145
.

.

%

*

*

.

'
Appendix A: Full Scale Testing Techniques

A . 1 Laboratory Related Problems and Logistics

A. 1.1 Introduction

During the planning of the project it was decided, following discussions with

CALTRANS engineers, that the effect of axial loads of approximately 0.09f' cAg

and 0.18 f' cAg would be investigated in the first test series. For a 5 ft.

(1.52 m) diameter column with 4000 psi (27.6 MPa) design concrete strength,

this corresponded to axial loads of 1000 and 2000 kip (4.45 and 8.9 MN) . The

National Bureau of Standards had at its disposal a 12,000 kip (53 MN) , manually controlled universal testing machine (Figure A.l) which could meet

these loading conditions. The machine was equipped with a tie-down floor

system (see Figure A. 2) which extended 42 feet (13 m) either side of

centerline. The floor system contained 126 bolt emplacements, each capable of

handling 150 kip (668 KN) loads in tension. The slab was designed to withstand

a maximum applied moment of 8400 kip-ft. (11.4 MN-m)

.

Ideally, laboratory tests should be conducted to reproduce loading conditions

anticipated in the as-built structure. Earlier it was mentioned that loading

of single-bent cantilevered bridge columns during seismic events is

predominately due to inertial loads from the bridge superstructure (both

lateral and vertical) and the gravity weight (static reactions) of the

superstructure. Because most of these bridges use elastomeric bearing pads at

the column/superstructure joint, the connection can be modelled as a hinge.

For the case of the 1/6 scale model tests, as shown in Figures A. 3 and A. 4,

it was possible to achieve these boundary conditions through use of the NBS

Tri - Directional Testing Facility (TTF)

,

a computer controlled loading system

147
A.l: Artist's Rendering of 12000 kip

Test Machine.

148
I.OMGITUIHNAI

111.

MiI

N(,

MOMKN'I

12,500,000

Ibf-ft

(

<'V

K

CKNTKAI.

'.()

H.I.I)

20,000,000

lbf-ft

7 bolts 2’ -0" o.

c

j

149

Figure

A.

2

12000

kip

Testing

Machine

and

Tie-Down

Floor
TTF TEST SET-UP

Upper (loading) Computer controlled

Figure A. 3: Lateral View of Model Specimen

Mounted in Triaxial Test Facility

150
Figure A. 4: Isometric View of Model Specimen and

Control Loads Applied by Triaxial Test

Facility. East West Actuator was used

only for Maintainance of Lateral Stability.

151
in which both axial and lateral loads could be applied to the top of a test

column by means of a moving loading head. Vertical alignment of the axial

load was thus assured and appropriate P-A effects were generated during

lateral deflections. The column "foundation", as shown in Figure A. 3, was

rigidly attached to the test floor, thus assuring a fixed-end condition at the

column base.

The Tri - Directional Test Facility was limited in use to relatively small

specimens. The 12000 kip (53 MN) testing machine, on the other hand, had one

primary drawback: the loading head could not sustain any significant lateral

displacements due to the likelyhood of rupturing its sensitive membrane load

cell. This problem is generally endemic to large capacity vertical testing

machines . Because of this several alternative loading methods were

investigated, all of which relied upon the technique of moving the column

base, while the column top -- which had a hinged boundary condition similar to

the model specimens -- was restricted from lateral displacements. One

approach used in bridge column tests in New Zealand [7] was to construct a

symmetrical specimen as shown in Figure A. 5. Lateral loads were applied to

the central block while the two ends were pinned [i.e. restrained from lateral

motion]. This had the potential benefit of obtaining two tests (the top and

bottom specimens were identical ) for the price of one, assuming that

interaction effects between the two specimens were minimal, and that failure

was nominally symmetric. A pronounced asymmetric failure (formation of a

hinge on either side of the central load block, but not the other) would have

resulted in minimal or no useful data in the inelastic regime from half of the

specimen. This was considered an unacceptable risk given the cost of a fullscale specimen. Regardless of the above considerations, use of a specimen

such as shown in Figure A. 5 would have resulted in full-scale specimens taller

than the 60 foot ( 18.3 m) vertical access space available inside the testing

machine. It would also have demanded the use of a substantially stiffer,

taller reaction system for resistance of lateral motion at the upper column

hinge. Consideration of these factors eventually led to the moving specimen

concept described in the next section.

152
oov

03 —

,




rQJor-“ 2

<1) 4-10> P4C

•H

>i

-Q

153
A. 1.2 The Moving Specimen Concept

Figure A. 6 and A. 7 show schematics of the test setup, specimen, and data

collection system employed for the full-scale tests. In this configuration

the column is rigidly cast into a massive, mobile foundation structure. The

depth of the base section was designed to accomodate longitudinal bar

development lengths as called for in present CALTRANS specifications. For the

columns tested in this program, the embedment depth was 8.5 feet (2.6m) to the

point of tangency of the bar hooks.

o The base was reinforced with 15-#11 (1.56 sq.in; 1006 mm^) grade 60 deformed

bars for compression and 30-#14 bars (2.25 sq.in.; 1452 mm^) grade 60 deformed

bars for tension, as shown in Figures 2.1 and 2.2. This reinforcing pattern

was sufficient to withstand the moment and shear forces developed at maximum

anticipated column plastic moment under a maximum axial load of 2000 kip with

a load factor of 1.5. Moment capacity of the base section was analyzed using

classical theory for doubly reinforced beams; shear capacity was determined by

means of a deep beam analysis using procedures proposed by Zutty [8]. The

worst case design criteria for moment in the base block occurred along the

vertical section flush with the column face on the side carrying the highest

roller support reaction. This corresponded, for a column sustaining an axial

load of 2000 kip (8.9 MN) and a lateral deflection sufficient to develop the

plastic moment of the column (approximately 8000 ft. -kip ; 10.9 MN-m) , to a design moment for the reaction wall of 21,000 kip-ft. (28.5 MN-m).

The final dimensions of the footing were 26 ft (7.92 m) long x 9 ft. 3 in.

(2.82 m) high x 8 ft. (2.43 m) wide. This base block and its monoiithically

cast column test section were supported by continuous track roller systems at

either end of the footing, as shown in Figure A. 6. The spacing between the

rollers and the mass of the base block were designed so that no uplifting at

either support would occur during the tests, and also to ensure that failure

would occur in the column rather than the footing. During the course of any

particular cycle the rollers on one side of the specimen were unloading

154
C/) >-

cn<

QC

O

CO

CM

155

Figure

A.

6

Schematic

of

Full

Scale

Test

Apparatus
DATA

CHANNELS

,

/

DATA

CHANNELS

LARGE SCALE STRUCTURAL RESEARCH FACILITY

SERVO CONTROL

Figure A. 7: Data Acquisition System and Test Control for

Full Scale Bridge Column Tests.

156
while those on the opposite side were picking up more of the vertical

reaction. The roller system, therefore, had to be designed such that any side

could handle the entire weight of the specimen plus the applied axial load.

Figure A. 8 shows two charts used to determine the required spacing of the

roller transport system. The first of these shows the minimum roller spacing

to prevent uplift on the unloading side of the base block as it is displaced

laterally from center position. A conservatively low axial load of 800 kip

(3.6 MN) was chosen for consideration of the uplift problem. The second chart

shows the maximum roller reaction as a function of spacing on the loading side

of the base block. The combined allowable load for the four roller units at

each end of the specimen was 2400 kip ( 10.7 MN) . The design moment was 9240

kip-ft. (12.5 MN-m) , although moments as high as 12,000 kip-ft. (16.2 MN-m)

were anticipated if the cured concrete strength proved substantially higher

than the 4000 psi (27 MPa) design strength. The final spacing (L) chosen was

23.4 ft. (7.1 m) which provided a factor of safety of 1.5 for uplift and 1.16

for crushing.

157
APPLIED

MOMENT

(1000

kip-ft)

APPLIED

MOMENT

(1000

kip

-ft)

MAXIMUM ROLLER LOAD (kips)

Figure A. 8 Design Chart for Roller Reaction

Forces as a function of base length.

158
A. 1.3 Rail Transport System

The key to the mobile specimen concept was a roller subsystem. This consisted

of eight 300 ton (2.7 MN) capacity, specially modified Hillman Rollers as

shown in Figure A. 9. These were constructed from a continuous chain of 20

hardened rollers, each measuring 2 in. ( 51 mm) diameter by 5 in. (130 mm)

long held in alignment by means of external connector links fastened at each

end of the rollers. The surrounding frame provides the reaction system which

carries the loads from the specimen to the individual rollers. Four smaller

roller bearings, oriented vertically, were affixed to the corners of each

roller system to keep them centered on the rails described below. Four of

these roller systems were positioned at each end of the specimen base block,

so that a total of eight, with a vertical load capacity of 2400 tons (21.4 MN)

were available for specimen handling purposes. Half of this, or 1200 tons

(10.7 MN) could be counted on for resisting the 1100 ton (9.8 MN) , worst case

vertical reaction resulting from the loading and unloading associated with

cyclic loading described above.

It is important to recognize the difficulties of handling a specimen weighing

190 tons (1.7 MN) in a laboratory environment. This was well beyond the

capacity of both laboratory overhead cranes (combined capacity 55 tons,

0.49 MN) and therefore once constructed alternative mechanisms of transporting

the specimen into the test apparatus were necessary. This led to the concept

of a laboratory rail system (Figure A. 10). The system consisted of four

separate tracks, each 120 feet (37 m) in length, spaced on two foot (610 mm)

centers, upon which the 300-ton roller units would ride. These were

fabricated from 8 in. wide by 2 in. high (203 x 51 mm ) sections of high

strength T-l steel (100 ksi , 690 MPa yield stress). Construction tolerances

allowed for a maximum deviation from level of 0.050 in. (1.3 mm) averaged

over the length of the track. This required the use of machined shim plates.

In addition to providing a means for cyclic loading of the specimen, the rail

system permitted easy access and removal of the specimen from the test

apparatus and allowed for construction -- and later demolition -- of the

specimen at a site some distance away from the test apparatus. Figure A. 11

shows a sketch of the transport system and test facility.

159
HKiH

bIHtNUlM

160

Figure

A.

9

Roller

Transport

Units

used

for

Specimen

Support
® *4 ®

w»

161

A.

10

Laboratory

Rail

System
NBS

LARGE

SCALE

SEISMIC

TEST

FACILITY

1

•a

<U —

162

Figure

A.

11

NBS

Large

Scale

Seismic

Test

Facility
Inherent with the use of the moving specimen concept is the understanding that

the roller transport system is not frictionless, and lateral load losses due

to friction between the rollers and the rail must be accounted for in order to

arrive at the actual applied column moments. The initial estimate of the

coefficient of friction for the rollers was 5% of the applied axial load,

.which could, in turn, have led to a substantial difference in the actual

lateral load perceived by the test column.

To accurately account for any losses that might arise from friction, a series

of cyclic load calibration tests were conducted on the rollers as shown in

Figure A. 12. Two of the roller systems were mounted in the 12000 kip (53 MN)

test machine such that a length of the T-l track used in the actual rail

system could be sandwiched between them. The track, in turn, was connected to

a servo -hydraulic actuator capable of moving the track back and forth

following a sinusoidal control pattern. A series of increasing axial loads

(perpendicular to the motion of the track) were applied to this assembly up to

a maximum load of 300 tons (2.7 MN) , the service capacity of the roller

systems. At each increment in axial load two cycles of lateral displacement

were carried out and the resulting lateral loads, displacements and axial

loads were recorded. A sample of the experimental load- displacement data is

presented in Figure .A. 13. The horizontal sections, which remain essentially

stable throughout the majority of the lateral displacement cycle, represent

the kinetic rolling friction resistance. The spikes near each end represent

the transition between kinetic and static rolling friction resistance. Since

a smooth, continuous motion was to be provided during actual testing of the

column specimens (by means of a servo -hydraulically controlled actuator) , and

since the static transition represented only a small percentage of the total

displacement cycle, a friction coefficient calibration curve was established

as a function of axial load using the constant kinetic friction data. The

resulting friction calibration curve is shown in Figure A. 14. This indicated

a linear variation of the rolling friction coefficient as a function of

applied axial load, with a peak valve of 0.0015 at the maximum individualroller service load of 300 tons (2.7 MN) . This was substantially less (more

than an order of magnitude) than the original 5% worst-case estimates provided

163
SSSS

Figure A. 12 Roller Friction Calibration

Test Apparatus

164
LATERAL

LOAD

(kips)

FRICTION CALIBRATION FOR HILMAN ROLLERS

•AB

-C

•D

-E

•F-G

•HJ

K

•L

Axial

Load

(kips)

100

150

200

250

300

350

400

450

500

550

600

Figure A. 13 Roller Friction sample Calibration

Data for varying Axial Load

165
450

o

LO o LO o LO o

Is- o 04 LO IsCO CO C\J

I—

Z

LU

g

LL

LL

LU

O

O

o

hg

QC

(sdpi) avoi 1VIXV

•5

4-1o

-M

23 •H

o

•H

4-1 ^ 4-1 T3

8 S

O rH

s> 13 EM

5 a

C ro o

•H 4-1

4-> O

03U C

3.

rH -P 5 §

Jj 'U

. 2 W

*2

ff-S

3ts r-1 -H

a m

<

166
by the manufacturer. For an axial load of 1000 kip (4.5 MN) , the value used

for both full scale tests, and with the specimen at maximum lateral

displacement from centerline this amounted to less than 0.25% of the applied

lateral load being dissipated in friction.

Despite the relatively low friction losses, the decision to monitor friction

forces was made upon initial acquisition of the roller system, owing to the 5%

friction estimates provided by the manufacturer. Given a calibration curve

it was possible to directly, determine friction forces provided the axial

loads being carried by each roller system ( eight of them) were known. A

series of custom designed load cells were developed as shown in Figures A. 15,

A. 16, and A. 17 to accomplish this purpose. These are described in greater

detail in section 2.4,2 under instrumentation.

A. 1.4 Reaction Wall System

At the outset of the project the existing apparatus in the NBS Large Scale

Test Facility consisted of the 12000 kip vertical actuator and its tie down

floor, as shown in Figures A.l and A. 2 The mechanism developed for resisting

the large lateral loads necessary for testing of the bridge columns is

described below.

Figure A. 18 shows the lateral reaction forces to be resisted during testing of

the full-scale 30-foot, and 15 foot (9.1 and 4.6 m) columns. For design

purposes it was decided that column axial loads as high as 3000 kip (13.4 MN)

would be considered, as would columns only 10 ft ( 3m) in height. Both of

these criteria imposed harsher requirements for the reaction wall than would

be required for any of the columns described in this report in order to

provide capacity for future expansion of test parameters. For nominal material

properties (grade 60 reinforcement, f' c = 4000 psi) this resulted in an

ultimate moment capacity for the 5 ft (1.52 m) diameter column previously

described of approximately 9240 kip-ft (12.5 MN-m) . The resulting maximum

overturning moment to be resisted by the lateral reaction system was therefore

167
LOAD

CELL

168
Figure A. 16 Close-up of instrumented Flatcell

prior to placement under specimen

base

.

169
1

1

1

I

Figure A. 17 Placement of Roller/Flatcell stacks underneathe

Full Scale Specimen.

170
o

CO

CO

CO

LLI cc

X =3 05 o

u.

O

171

A.

18

Lateral

Reaction

Forces

Present

Curing

Full

Scale

Testing
10,830 kip-ft (14.7 MN-m) for a 30 ft (9.1m) tall column and 13998 kip-ft.

(19 MN-m) for a 10 ft. (3m) tall column as shown in Figure A. 18. The maximum

shear force occurred during loading of the shortest column (10 ft., 3m) and

was equal to 924 kip (4.1 MN)

Several competing designs for the reaction system were conducted in parallel

and are shown in Figure A. 19. These were a monolithically cast, vertically

post- tensioned reinforced concrete "H Beam"; a segmentally precast, vertically

post- tensioned reinforced concrete "H Beam"; and a braced double-bent steel

"V-Frame". The steel option was eliminated on the grounds of excessive cost,

and flexibility: while strength criteria was easily met, deflections at the

upper tie-back location for the 30 ft. (9.1 m) column were 0.32 in. (8 mm)

which was considered unacceptable due to the previously described problem of

permitting lateral loads to be carried by the vertical test machine. The

monolithically cast H-Beam was more cost-effective than the segmental route

and deflections were within tolerance, provided the reaction floor system was

presumed to act as a rigid foundation.

A. 1.5 Deflection Considerations

Besides the sensitivity of the vertical test machine to local lateral

deflections (i.e. the reaction wall displacement under load with respect to

the column top)

a further design consideration was the proximity of the

National Deadweight Measurement facility located 80 ft (24 m) from the site of

the Large Scale Test Facility. Since both laboratories were housed under the

same roof, and built on the same foundation, concern was raised as to the

possible detrimental effects of large deflections that might be introduced

into the building foundation during the course of a bridge column test. This

question was addressed as follows: borings were made at three locations in the

vicinity of the testing machine and the proposed reaction wall to bedrock at a

depth of 90 ft. (28 m) . Material properties were then calculated for each

distinct soil layer and a detailed two dimensional linear elastic finite

element analysis (Figure A. 20) of the test facility was carried out which

included the underlying soil stratifications. This analysis showed excessive

172
W14x99

STEEL REACTION FRAME

Total weight = 82 tons

1 I I I I

SEGMENTAL PRECAST REACTION WALL

Mondothic wall similar but eleminates

horizontal joints and shear keys

Figure A. 19 Competing Reaction Wall Designs
» v- mm

Figure A. 20 Finite Element Model of

Wall and Tie-down Floor

Reaction

System

174
rigid-body rotation of the tie-down floor system due to compression of the

underlying soft clay sediments . The model was verified by a subsequent dead

load check (stacking of approximately 100 kip [450 KN] of mass at the location

of the reaction wall) during which floor displacements and rotations were

monitored. The final reaction wall design (Figure A. 21) therefore made use

of two 42 in. (1.1 m) drilled caissons which anchored the wall to bedrock.

This approach limited lateral deflections at the top of the reaction wall to

less than 0.1 in. (2.5 mm) under test load conditions.

Connection of the reaction wall to the existing tie-down floor was achieved by

means of 23 post- tensioning tendons (see Figure A. 22), each carrying a preload of 150 kip (668 KN) . These, combined with the dead load of the wall

produced an axial precompression stress of 0.206 ksi (1.42 MPa), with a resultant decompression moment of 18660 kip-ft. (25.3 MN-m) and a shear

capacity of 1595 kip (7.1 MN) . These produced load factors of 1.7 for both

moment and shear for the design loads described -earlier

.

A. 1.6 Tie-Back Frame

Figure A. 23 shows the tie-back "A-Frame" used to carry lateral reaction loads

from the column top to the reaction wall. The frame was designed to be

hinged at both ends so as to follow changes in the top - of - column elevation

during cyclic lateral loading. It was furthermore designed to resist any

torque generated about the column axis due to eccentric failure (spalling

etc.) of the column during testing. Service loads for the A-Frame were 1200

kip (5.3 MN) in tension and compression with applied in-plane moments due to

torque of 400 kip-ft (0.54 MN-m). This was connected to the reaction wall by

means of -three positioning beams (Figure A. 24) which were in turn posttensioned to the reaction wall. The positioning beams permitted the

attachment of the A-Frame to the wall at any height; without them attachment

points were possible only at fixed heights at 5 ft. (1.52 m) intervals.

Strain gage instrumentation was applied to the A-Frame at the two locations

shown in Figure A. 23 so as to provide two full-bridge loadcells at each

175
a -

2 Ui a

-

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< O §

- ; sa2s<

5 5 o i

176

figure

A.^l

filial

Reaction

Wdii

Design
Figure A. 22 Vertical Post- tensioning of Reaction

Wall to Tie-down Floor

177
I

I

178
179

Figure

A.

24

Vertical

Positioning

Beams

used

for

attachment

of

Test

Apparatus

to

Reaction

Wall
location. The A-Frame was calibrated to 1000 kip (4.5 MN) using the 12000 kip

(53.4 MN) testing machine prior to each full-scale test. This provided a direct measurement of losses in the applied lateral load due to friction in

the test system, and also a measure of any torque caused by eccentric failure

of a specimen.

A. 1.7 Full-Scale Hinge Boundary Condition

For similitude to exist between model and prototype, the top of the full-scale

column required a hinged boundary condition. For the model columns it was

possible to achieve this by using a hardened cylindrical rod bearing on two

hardened plates. This option could not be used for the full-scale columns

due to safety considerations surrounding the possibility of a falling rod

weighing several tons. An integral, one piece hinge was designed and

constructed as shown in Figure A. 25. The top surface of the hinge (that which

would be in contact with the vertical loading machine head) was faced with a 1

in. (25 mm) plate of hardened 4340 steel. During the second test (aspect

ratio = 3) some 600 - 1.5 in. (38 mm) ball bearings, placed in molybdenum

disulphide grease in a hexagonal closed packed pattern and capped with a second 1 in. (25 mm) 4340 steel plate, were added to the top of the hinge.

This served to reduce transmittal of lateral force to the vertical test

machine, which was substantially stiffer for the shorter column height of the

second test. Figure A. 26 shows this hinge being attached atop the second

full-scale test specimen. Mortar was used to seat and level the hinge on top

of the column prior to vertical load application. The hinge assembly weighed

12 kip (53 MN)

The use of the hinge and tie-back frame as described, while satisfying the

required top-of -column hinged boundary condition for the full-scale tests, did

not precisely reproduce the model loading situation. The reason for this is

as follows: in the model tests, the lateral load was transferred to the column

by means of a hinged rod bolted to the column cap block, in a manner similar

to that for the prototype specimens. As with the prototype specimens, the

top - of - column hinge for the model specimens was not coincident with the

lateral tie-back point. However, because both the vertical and lateral loads

180
4340

HARDENED

BEARING

PLATES

i

600

I

1/2'

BALL

BEARINGS

(Shear

Column

Only)

181

A.

25

Full

Scale

Hinge

Boundary

Conditions
I

; Figure A. 26 Full Scale Hinge being positioned on

top of Shear Column Specimen

182
for the model were applied using a crosshead which moved with the column top,

negligible lateral force was transmitted to the column at the level of the top

hinge. This was not the case with the full scale tests, since it was the

bottom of the column that was moving. The top hinge, therefore, moved

laterally in proportion to the cyclic lateral displacements imposed at the

bottom of the column and generated potentially significant lateral forces at

the top hinge level due to the bending stiffness of the test machine. This

lateral machine reaction acted counter to the direction of the tie-back

reaction force (see Figure A. 34) and explains the need, as described in

section A. 6.

2

to measure the force dissipated into the test machine in order

to accurately determine column moment.

An alternative hinge design (Figure A. 27), in which the axes of both the tieback frame connection and the top - of - column hinge coincided, was considered in

the early design phase of the project. While this would have effectively

eliminated the need for instrumentation of top-of-column hinge displacement

the fabrication costs were considered prohibitive at the time. Further

discussion of this detail is presented in Appendix B. The mechanics of the

as-built test apparatus are discussed in Section A. 2.

A. 1.8 Apparatus Connections

Figure A. 28 shows the A-Frame attachment to the specimen. All such

connections, including those between the specimen base and the lateral load

actuator, were accomplished with the use of eight Dywidag 1-3/8 in. (35 mm)

threaded post- tensioning bars (Figure A. 29), each of which carried a preload

of 110 kip (490 KN) . This, together with similar post- tensioned connections

to the reaction wall insured that there would be no slack in the system at

load reversal during cyclic lateral loading. Figure A. 30 shows a view of the

completed lateral reaction system prior to testing.

183
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184

A.

2/

Alternate

coincident

Hinge

Considered

tor

Rill

Scale

Test

Apparatus
Figure A. 28 A-Frame attachment to Column Specimen

185
I

Figure A. 29 Lateral Post-Tensioning of

A-Frame assembly to column top.

186
Figure A. 30 The completed Lateral Reaction System

Prior to Testing showing upper and lower

Vertical Positioning beams, Tie-back

Frame, Lateral Actuator, and Reaction Wall.

187
A. 1.9 Loading Systems

Vertical load was applied using a 12000 kip (53 MN) universal testing machine,

as shown in Figure A. 11. Because this was to be used to simulate the gravity

load of the bridge superstructure, it was necessary to maintain constant load

throughout the test. The testing machine was originally designed with manual

operation in mind, similar to most other universal testing machines. Accurate

load control, therefore, would have been a difficult task during cyclic

lateral loading of the test specimen, due to the change in elevation at the

top of the column as rotation took place at the column base. A programmable

servo-valve was installed within the hydraulic power supply system to the

loading cylinder and was found to have stable response and good tracking

ability when driven by a programmed control signal. For both full scale tests

the vertical test machine was run under load control with a constant applied

axial force of 1000 kip (4.5 MN) . Maximum local vertical displacement range

[ ram stroke] was 24 in. (0.61 m) prior to the point where the load crosshead

would have to be relocated using the exterior screws.

Lateral load was applied using a 1000 kip (4.5 MN) servo - controlled

programmable double acting hydraulic jack as shown in Figure A. 31. This had a

maximum stroke of 48 in. (1.2 m) for monotonic loading. For cyclic loading

the maximum stroke either side of null was 24 inches (0.6 m) . Design

calculations indicated that this displacement would be sufficient to reach a minimum of 7 Ay for the 30-ft. (9.1 m) column and 14 Ay for the 15-ft. (4.6 m)

column, thereby ensuring the ability to completely fail both columns. During

a full-scale test a "ramp and hold" load- time response was programmed into the

servo-controller so that the specimen moved from its central null point to a

specified multiple of yield displacement in a continuous, smooth motion. The

exception to this procedure, as will was described section 3.1, was the

determination of yield displacement. For that particular case, load control

was used to reach a specified lateral force.

A plan view of the completed Large Scale Seismic Test Facility is shown in

Figure A. 32.

188
Figure A. 31 Servo-controlled 1000 kip hydraulic

ram used for Lateral Load. Also note

Lifting Bracket and Corner Jack at

Lower Right Hand Corner of Base Block

for Raising and Lowering Specimen.

189
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1-2

Figure A. 33 Mechanics of the Test Apparatus

191
A. 2 Mechanics of the Test Apparatus

The complete test apparatus, as shown in Figure A. 11 constitutes a statically

indeterminate system, as described in Figure A. 33. Without special procedures

the system consists of 18 unknown reactions and three equations of

equilibrium. The known quantities are the applied lateral load and the

applied vertical load, which are measured by means of calibrated load cells.

The base column moment and shear can be determined independently by two

different means

Using top-of -column reactions:

Ml = P af (h 1 + Ah )

- Pm

(h 2

) + P a (A t ) + W i A(h)dh (A.l)

but since Ah is very small and the moment caused by the displaced

weight of the column can be approximated by the column weight times

the lateral displacement of the column centroid, Eq . (A.l) can be

re-written as:

Mi - P af (h!) . Pm

(h 2 ) +P a (A t ) + Ui h(AT/ 2 ) (A. 2)

/l = P a f Pm ( A . 3

Using foundation reactions:

4

M2 - 2 (L2 )/ 2

i=l

8

2

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8

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i=l

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8

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r

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i=l

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4

2

i=l

N i(L2 )/2

8

2

i=5

Ni(L2 )/2

(A. 4)

(A. 5)

(A. 6)

Mi = M2 = M and V L = V2 Note that: = V

192
In the above equations (as well as Figure A. 33 ) the following definitions

apply:

Mq = Column base moment as calculated

from top of column reactions (kip- in.)

M2 = Column base moment as calculated

from foundation reactions (kip-in.)

Vq = Column base shear as calculated

from top of column reactions (kip)

V2 = Column base shear as calculated

from top of column reactions (kip)

/iq = Coefficient of friction at i(th) roller support

= Vertical offset of tie-back hinge

due to top - of - column rotation (in.)

Ar = Lateral deflection at base of column (in.)

A t = Total displacement for computation of

P-A moment ( in.

)

V = Column base shear (kip)

Km = Vertical test machine bending

stiffness (kip/in.)

Fq = Friction force at i(th) roller support (kip)

Nq = Vertical reaction at i(th) roller support (kip)

M = Column base moment (kip-in.)

dh = integration parameter for moment due to column weight,

where h is the height above the column base,

ho = Height of specimen base block (in.)

hq = Effective height of test column (in.)

h2 = Height of top - of - column hinge axis (in.)

hq = h2 - hq ( in.

)

h^ = Vertical test machine crosshead height (in.)

W =* Specimen weight (including base block) (kip)

P a = Axial load (kip)

P a f= Reaction wall load (kip)

Pm

= Load dissipated into vertical test machine (kip)

P r = Lateral actuator load (kip)

Lq =» Lateral offset for tie-back frame hinge (in.)

L2 = Center- to -center distance between

roller supports (in.)

Wq = Unit weight of column (kip/in.)

A(h) = Lateral deflection of column (3 height "h" (in.)

The unknowns are the vertical reaction forces at each of the eight roller

supports, the friction forces developed at each of the 8 roller supports, the

lateral load resisted by the tie-back frame, and horizontal forces dissipated

into the vertical testing machine through the hinge assembly. Procedures were

taken to experimentally measure all of these unknowns.

193
The vertical reactions at the eight roller units, as well as friction forces

developed at each roller, were to be monitored by calibrated, specially

fabricated "flat cells" -- low aspect ratio machined veerendeel trusses

instrumented with strain gages and calibrated to measure both axial and

shearing forces. These are described in section A. 6.

(instrumentation) in

greater detail. The previously described tie-back frame was instrumented to

form four full-bridge strain gage loadcells (see section A. 6. 3) so that both

lateral load and column torque could be monitored. This was calibrated prior

to each test by means of axial loading with the 12000 kip (53 MN) vertical

test machine. Lateral deflections at the test macine loading head were

monitored using a rotary potentiometer mounted to a stable reference structure

30 ft. (9.1m) distant from the test apparatus and connected by means of light

gage stainless piano wire. Provided an accurate lateral stiffness could be

determined for the test machine at each test elevation, crosshead displacement

could then be used to directly determine the lateral reaction transmitted to

hinge at the top of the column. Such stiffness calibrations were performed

for each full-scale specimen as shown in Figures A. 34 and A. 35.

It should be noted that the actual moment in the test column at any given load

stage could be determined within specified bounds of error by completely

knowing either the reaction forces at the top of the column, or those at

acting on the base structure. The instrumentation of the tie-back frame and

the determination of machine lateral bending stiffness therefore provided a

redundant mechanism for determination of losses in applied lateral load due to

friction in the continuous track roller support system. Furthermore, the

eight unknown friction forces could also be determined by knowing the vertical

reactions at each roller system and making use of the calibration curves

relating coefficient of friction to axial load. The effectiveness of the

various techniques used to monitor reaction forces is discussed in Chapter 6

as well as Section A. 6 .

194
A. 3 Access Techniques for Large Scale Tests

When designing the Large Scale Test Facility, as well as the prototype column

specimens, there was an early appreciation of the need for rapid access to

significant elevations -- in excess of 50 ft. (15 m) above the test floor in

some cases. This generally precluded the use of ladders and off-the-shelf

scaffolding. Furthermore, specially constructed permanent stair systems and

catwalks would inhibit viewing of the specimens under test, given the space

restrictions of the facility, as well as impede placement and removal of test

apparatus, since speed of assembly and dissassembly was directly related to

ease of overhead crane access to the heavy test setup components.

A somewhat novel approach was adopted as a solution to the access problem.

This involved the use of "single rope techniques" [References 8 and 9] in

which a researcher could ascend or descend fixed nylon ropes suspended either

from the reaction wall, the test machine, or laboratory roof trusses (Figure

A. 36). This had the distinct advantages of spontaneity -- in minutes a lightweight line could be rigged wherever needed -- and lack of clutter that

would be associated with scaffolds, ladders and other more traditional

methods

.

The method employed a relatively stiff 7/16" (11 mm) kernmantle nylon rope

with a braided external sheath (which served to resist abrasion) and a straight- lay , multifiber core. Breaking strength of the line was 7700 lb.

(34 KN) and at any given time during the conduct of a full scale test, some

400 ft. (122 m) were rigged at various key locations throughout the laboratory.

Researchers and technicians wore alpine style seat harnesses and were able to

ascend these lines using mechanical ascenders attached to the harness by means

of 1 in. (25 mm) tubular nylon sling which had a breaking strength of 4400 lb.

(20 KN) . Descent was accomplished by means of a variable friction "rappel

rack", as shown clipped to the harness of the researcher (Figure A. 37)

marking cracks on the full-scale flexure column. A total of four members of

the laboratory staff ( none of whom had any prior experience in the use of

such hardware) received professional instruction. All learned quickly and

195
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Figure A, 36: Vertical Access to Difficult Places

was Achieved Using Technical Ropework

Methods

198
I

li

Figure A. 37: Crack Marking on Full Scale Flexure

Column Test Specimen.

199
were able to safely apply the new skills without incident. Ropes were

replaced regularly on a 6-month basis and all slings were inspected prior to

each use and replaced without hesitation at the first signs of wear.

One particularly useful application of the above described access techniques

was that of crack-marking (Figure A. 37). Here several lines were suspended

from the vertical loading head thus permitting rapid access to the entire

surface area of the column while not interferring , with photographic

documentation. Felt pens typically used for marking cracks in model

specimens were not effective for full scale tests since they were too thin for

reasonable resolution when viewed at the distances necessary to photograph the

entire column (approximately 30 feet, 9.1 m) . Instead, 0.5 in. (12 mm) foam

tipped glue applicators filled with drawing ink were used as shown in Figure

A. 37. Another alternative which was considered, but in the end not used

because the first approach -- the ink applicators -- worked so well, was that

of using variously colored spray paints equipped with extension tubes for

applying a narrow band. This would have permitted the use of separate colors

for identification of the various lateral displacement levels.

While the benefits of single rope techniques for full scale testing (both in

the laboratory and in the field) are clear, so is the need for expert

instruction for those researchers contemplating the use of such techniques.

The authors recommend reading Reference 10 and then contacting a respected

alpine or speleological society for sources of training.

A. 4 Specimen Demolition and Removal

Following testing of a ^full-scale column there was the significant task of

disposing of the now-useless specimen. The use of explosives was considered

the most effective method but could not be used indoors owing to the proximity

of sensitive test equipment. The method eventually employed was as follows:

1) Prior to casting of the base, insert three horizontally oriented, 2 in.

(50 mm) diameter thin walled steel conduits equally spaced across the

height of the base block at each of the third-points along the length of

200
the base block. Ensure that these are not plugged during concreting

operations

.

2) Prior to casting of the base, install 0.625 in. (15 mm) diameter thin

walled aluminum conduits oriented perpendicular to the side (long) face of

the base block and lying flush with each layer of longitudinal deformed

reinforcing. There were three such layers: a top layer consisting of 15

#11 bars and two bottom layers consisting of 15 #14 bars. Three such

conduits were placed at each of the third points along the length of the

base, flush with the vertical tubes described in 1) above.

3) Prior to casting of the base, install concrete lifting anchors (see

Figure A. 38) such that, once demolition operations have been completed, the

base can be hoisted in three separate, approximately equal segments. Each

segment must have sufficient embedment anchors to permit stable hoisting

and be relatively insensitive to minor errors in the location of the

fracture planes generated during the demolition process.

4) Following completion of the test, burn through all longitudinal column

reinforcement at the column base using a torch. This is generally a rapid

procedure since most reinforcement is either buckled or fractured at this

stage already. Severing of the final reinforcement should not be attempted

until the column has been rigged to an overhead crane. With the crane in

place and the final bars removed, the column can be laid down horizontally

on the base block at which point a second crane can be attached at the

opposite end for loading to a transport vehicle. If a large capacity

overhead crane is available (50 ton) this operation can be done in one

motion. Figure A. 39 shows a typical full-scale column ready for transport

from the testing lab.

201
Figure A. 38: Installation of Cast- in-place lift

Anchors to be used for Subsequent

Demolition of Specimen.

202
5) With the column removed, fill the horizontal conduits with expansive

cement. A commercial grade known as "Bristar" was used with excellant

results. This cured within 24 hours and initiated a planar vertical

crack across the base block at each of the third points.

6) Using an oxygen lance, burn through the horizontal conduits. This is

a sufficiently vigorous process that the longitudinal deformed bars are

also consumed along the line of the conduit. Short oxygen lance burns

are also made along the side faces of the base at each of the fracture

lines to cut through the temperature and shrinkage steel.

7) Using jackhammers, cut a blockout in the top face of the base which

spans the fracture plane. Three of these are used across the width of

the block at each of the third point fractures and they must be

sufficiently deep to permit placement of large capacity hydraulic jacks -

-- at least 100 ton (890 KN) capacity each. These are then driven in

parallel to their maximum extension, resulting in the splitting- away of

the various segments. An oxy- acetylene torch is used to clear away any

residual steel and the segments are hoisted onto a waiting transport.

This was a time consuming, labor intensive solution but it was safely carried

out within the confines of a laboratory where several other test projects were

being conducted at the same time. Two alternative, and likely much less

costly, procedures for specimen removal are the above mentioned use of

explosives -- if the specimen can be initially transported to an outside

location following completion of a test -- and an approach involving different

specimen construction procedures which will be described in Appendix B.

A. 5 Material and Construction Process

A. 5.1 MATERIAL

The nominal size of the coarse aggregate used for the full-scale columns was

3/4 in. (1.9 cm). The specific gravity of the coarse aggregate in the

saturated surface dry condition (SSD) was 2.59. The fine aggregate had a

203
*

*

Figure A„39: Flexure Column Ready for Transport

from the Laboratory after Testing.

Maximum segment weight was limited to

60 tons (530 KN)

204
specific gravity SSD of 2.62, a specific gravity dry of 2.59 and a fineness

modulus equal to 2.57. The gradations of the coarse aggregate and the fine

aggregate are presented in Tables A.l and A. 2. The mix design for a concrete

compressive strength of 4000 psi (27.6 MPa), for 1 cubic yard - dry weights

SSD, are given in Table A. 3.

A. 5.

2

Construction Process

Commercial contracts were let for the tying of the steel cages and the casting

of the columns. Fig. A. 40a shows the steel for the shear column and some of

the formwork around the base. The formwork for the footing and the column was

made of steel for the flexure specimen. Wooden formwork was used for the

footing and steel formwork was used for the column of the shear specimen.

The inside of the base in shown in Fig. A. 40b. The wooden block-out on the

left foreground of the figure is the attachment point for the servo -hydraulic

ram applying the lateral load. This inset allowed the ram to be at its midstroke in the neutral or zero position and to have a maximum stroke of 24 in.

(610 mm) in the east and west direction. The pipes originating from the

block-out to the opposite end of the footing were used as conduits for the

post- tensioning bars attaching the ram to the footing. A closeup of the

confining reinforcement for the shear specimen prior to casting of the column

section is shown in Figure A. 41. Sprial reinforcement is #6 deformed grade 60

bar at 2-1/8 in (54 mm) spacing.

Due to the size of the column, application of strain gage instrumentation was

not easy. This was particularly true for the flexure column as climbing gear

was necessary to scale the outside of the column and to descend inside the

column to enable gaging of the longitudinal bars (Figure A. 42). Figure A. 43

shows part of the waterproofing process required for the gages on the

longitudinal bars. A description of strain gage locations and application

procedures is used is presented in Section A. 6.

205
Table A.l COARSE AGGREGATE

SIZE % PASSING

1 in. mesh size 100.0

3/4 in. mesh size 96.7

3/8 in. mesh size 49.5

#4 mesh size 6.4

#8 mesh size 3.2

Table A.

FINE AGGREGATE

SIZE % PASSING

3/8 in. mesh s ize 100.0

# 4 mesh size 98.1

# 8 mesh size 87.5

# 16 mesh size 74.9

# 30 mesh size 55.4

# 50 mesh size 24.7

# 100 mesh size 1.5

Table A.

MATERIALS WT. . LB

Cement, Portland ( 6 - 1/4 sacks), Type I 588.0

Sand 1100.0

Gravel 1936.0

Water (31 gal.) 258,23

Air Entrained

Water/Cement Ratio

Slump

5 %

0.44

4 in. + 1 in.

206
207

Figure

A.

40a

Full

Scale

Flexure

Column

prior

Figure

A.

40b

Inside

detail

of

column

base

to

casting

of

base

block

prior

to

casting
Figure A. 41: #6 Deformed Grade 60 Spiral Reinforcement

at 2-1/8 in. (54mm) pitch prior to Casting

of Full-Scale Shear Column Section. Scale

in inches

208
The columns were cast in two phases. The footing was poured in the first

phase and the column itself was poured in the second. This created a cold

joint at the interface of the column and footing which is standard

construction practice. Casting of the flexure column footing is shown in Fig.

A. 44- where the concrete is being pumped into the footing from a concrete

truck. Instead of pumping the concrete for the shear specimen, the concrete

was poured into hoppers and by using the overhead cranes, the hoppers were

lifted over the specimen where they were emptied as shown in Fig. A. 45. This

change in procedure was an effort to economize.

After removal of the forms lifting brackets (visible near the bottom right

corner of the base block in Figure A. 31) were attached at each of the four

corners of the specimen. Four 100 ton (890 KN) hydraulic jacks operated in

parallel were then inserted at the bracket locations and the entire specimen

was lifted approximately 12 in. (30cm) whereupon the bottom formwork was

removed, the railroad tracks cleaned, and one of the continuous track rollers

was placed under the specimen at each corner. The specimen was then moved

towards the test apparatus until access from the high-bay crane was possible

where the upper hinge assembly was grouted in place. Once in the test

assembly the lateral ram was connected to the base block and post- tensioning

operations were carried out. With the vertical loading head moved well above

the top of the column, the tie-back frame -- which was already attached to the

reaction wall -- was swung down in preparation for connection with the load

block at the top of the column. The specimen was positioned during this

operation with the use of the lateral ram under servo-control. The tie -back

frame was then post- tensioned to the top of the column and the vertical

loading head brought down into contact with the hinge assembly.

A. 6 Instrumentation

All columns were instrumented extensively in the potential plastic hinge

region. A total of 121 channels were used to acquire data for each test. Of

these, two were dedicated to record the lateral load and two channels to

209
Figure A.42 Vertical Access to Inside of Full-Scale

Column Reinforcing Cage Using Technical

Ropework Methods. Lines Circling the

Researcher Comprise the Wiring Harness

for first 50 Spiral and Longitidunal

Reinforcing Strain Gages.

210
Figure A. 43: Application of Strain Gage Waterproofing

to Longitudinal Reinforcing Steel

Instrumentation. Surface was Prepped

with a Carborundum Grinder, Followed

by Filing, and Sanding with Emory Cloth

Prior to Cleaning for Strain Gage Installation.

211
Figure A. 44: Casting Full-Scale Flexure Column Base.

212
Figure A. 45: Casting Full-Scale Shear Column

213
record the axial load. Seven channels were used to measure the column

displacement at various locations along the height of the columns using highresolution rotary potentiometers connected to a stable tower (Figure A. 46) by

means of 0.012 in. (0.3 mm) stainless wire. The rigging tower was located 30

ft. (9.1 m) distant from the test stand to eliminate the effects of local

displacements that would arise if the reference tower was mounted on the

specimen (due to flexural deformation of the base block) . During the first

(flexure) full-scale test two additional channels were used to measure the

rotation at the base of the column using linear variable differential

transformers (LVDTs) . The remaining channels were used to measure the strains

along the longitudinal and spiral reinforcement using resistance type foil

strain gages.

Redundant strain gaging was used in all critical locations, since repeat tests

would not be possible. Figures A. 47 and A. 48 show the locations of the strain

gages for the flexure column. From the results of the model tests and the

first full scale test, the strain gage and LVDT locations were modified to

better capture column behavior in the shear column. One such modification was

to place the spiral gages along the vertical lines of maximum flexural fiber

strain, the same as for the longitudinal bar gages. Previously, spiral gages

were located along vertical lines rotated 90 degrees from the longitudinal bar

gages to eliminate confusion when wiring instrumentation. It was felt that

this new orientation would allow for a better capture the first yielding of

the spiral.

A . 6 .Clip Gages

The two LVDTs used to measure column base rotation for the flexure specimen

were not used for the shear column test. In the interim between the first

and second full-scale test a new, and extremely sensitive "clip gage" was

developed to measure local strain (and hence curvature) throughout and beyond

the height of the potential plastic hinge region. The original concept for

this gage came from a similar device which had been developed at the National

Bureau of Standards for cryogenic fracture mechanics studies. The clip gages

214
it)

Figure A. 46: Stable Tower used for Column Displacement

Reference Plane.

215
were machined on a numerically controlled mill from 7075-T6 aluminum alloy and

had a gage length of 3 in. (76 mm) . An isometric drawing of the clip gage is

shown in Figure A. 49. A total of nine' clip gages were mounted on each

opposing column face along the line of maximum bending fiber strain. Anchors

for mounting the clip gages were tack-welded to the longitudinal reinforcement

as shown in Figure A. 50 at 15 in. (381 mm) spacing along the lower half of the

column. While this technique worked and did not appear to influence the

behavior of the longitudinal reinforcement, it was decided following the test

that a better procedure would have been to mount threaded connectors for the

clip gages on the ends of a continuous rod which extended through the width of

the column, thus providing solid support for each pair despite concrete

spalling etc. This procedure was used in the New Zealand tests [15]. Taken

together, the clip gages provided an accurate external measure of column

curvature along the height of the column. Experimental curvature at any

discrete height containing a clip gage pair was calculated as:

R = (eq^ - € 2i)/^ radians/in. (A. 7)

where = tension clip gage displacement/gage length = tension strain at elevation "i"

€ 2i = comP ress i- on clip gage displacement/gage length = compressive strain at elevation "i"

D c = out-to-out of spiral = 56 inches (1.52m)

A typical experimental calibration curve and best fit equation for clip gage

displacement versus output voltage is presented in Figure A. 51 which shows

excellent linearity over a 0.9 in. (23 mm) extension range. During actual

testing the clip gages were pre -compressed by 1/2 in. (12 mm) to accomodate

both positive and negative curvatures. Gage locations are shown in Figure

A. 52.

A. 6.2 Flatcells and the Measurement of Base Reaction Forces

The need to monitor friction and vertical reaction forces generated at the

roller supports was recognized early in the test program and a study was

216
1000 K

5

Figure A. 47: Location of Strain Gage Instrumentation

for Full-Scale Flexure Column.

217
1000 K

E- W

Figure A. 48: Location of Strain Gage Instrumentation

for Full-Scale Shear Column.

218
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Figure

A.

49

Isometric

View

of

clip

Gages

used

to

Measure

Column

Curvature
Figure A. 50: Clip Gages being Mounted on East Face of

Full-Scale Shear Column Specimen.

220
0006 0 I 221
Locations of Clip Gages and Rotary Potentiometers

for the Full-Scale Shear Column

1000 K

Figure A. 52: Clip Gage Locations for Shear Column Test

222
conducted to determine the methods available for such measurements. The

initial design conditions were that each of the eight roller systems would

need to be instrumented since final load distribution to each roller during

the course of a test was not predictable; that the friction forces could be as

high as 5%, and that service axial loads for each loadcell would be the 300

ton (2.7 MN) rated capacity of the rollers. As a result of the 5% friction

assumption (which was the only number available at the time, since the

coefficient of friction tests described in Section A. 1.3 had not yet been

conducted) each loadcell would have to handle maximum shearing loads of 30 kip

(133 KN) , perpendicular to the primary instrumentation axis. A survey was

made of existing commercial loadcells, including standard industrial axial

cells, and low aspect "doughnut" cells, and it was concluded that such

classical designs were not tolerant of high lateral load conditions.

Furthermore, none could serve as shearing force transducers. The decision was

thus made to pursue development of an in-house transducer which had high axial

load capacity, high shear force capacity, low aspect ratio, and high

sensitivity. The result is shown in Figures A. 15 and A. 16. This "flatcell"

was designed as a monolithically milled vereendeel truss. Strain sensitivity

was optimized through the use of a three dimensional finite element model and

the resulting dimensions were 31 x 12 x 2 in. (770 x 300 x 50 mm) . Machining

limitations controlled the final design and the maximum achievable width was 6

in. (150 mm) width; hence two such units were needed for each roller system,

with each cell rated at 300 kip (1.3 MN) service capacity. Sixteen such

devices were fabricated from 6061 T6 aluminum. Each unit weighed 17 lb.

(7.7 kg).

Each flatcell was instrumented as shown in Figure A. 53. Two configurations

were used. During the first full-scale (flexure) test web sections 2,6, and

10 (see Figure A. 53) were instrumented at mid-height with flexure compensating

strain gage pairs to monitor axial load. These six gages were connected in

series to average the load along the length of the cell. Three additional

independent strain gage pairs were mounted (see Figure A. 53b) on the same weD

sections near the flange to pick up friction forces. This gave a total of 4

instrumentation channels per load cell. It was planned to average the three

friction readings during an actual column test. By the time the first two

223
1 2 34 56 7 89 10 11 = Web Number

a) WIRING DIAGRAM, SECOND FULL-SCALE TEST

1 23456789 10 11

b) WIRING DIAGRAM, FIRST FULL-SCALE TEST

Figure A. 53: Wiring Schematic (two versions) for Flatcells.

224
prototype cells had been fabricated a friction test apparatus had been

assembled (Figure A. 12). Output response for the axial channels was linear

with applied axial load, and insensitive to lateral displacement. But the

measured friction- induced shearing strain did not correlate with the applied

lateral load. In view of the accurate calibration curves which were obtained

relating coefficient of friction with applied axial load a decision was made

to proceed with the first (flexure) full-scale test with only the use of the

axial load channels.

Once calibrated, a roller- loadcell stack was constructed as shown in Figure

A. 17, in which the loadcell was sandwiched between two ground (planar) 0.5 in.

(12 mm) plates and placed on top of a 1 in. (25 mm) elastomeric bridge bearing

pad which was seated on top of the roller system. Each of the eight rollerloadcell stacks were covered with a sheet of plastic. Hydrostone grout was

placed on top of the plastic and all eight rollers were positioned beneath the

test specimen using preset stops. The specimen, which was being held aloft by

means of hydraulic jacks at each corner, was then lowered until sufficient

contact was made to extrude the grout completely over each top plate. The

specimen was then locked at this position until the grout set, under the

assumption that any surface irregularities would be eliminated and that each

loadcell would see approximately equal load once the hydraulic supports were

removed. This assumption proved to be substantially in error.

During the first full-scale test electronic problems developed with two of the

axial load monitoring channels, one at each end of the specimen. This,

unfortunately, meant no empirical determination of the vertical reaction

forces nor the friction forces, since the reactions at all roller contacts

were required. Column moment was therefore determined from top-of -column

reactions

.

It was felt, following examination of loadcell data from the first full-scale

test prior to loss of the two chanels described above, that the use of a series gaging system was too prone to failure, since local yielding of one web

section (perhaps due to surface irregularities not compensated for with the

grout) might lead to failure of a gage, or might not be picked up at all. A

225
revised, semi - redundant gaging technique was used for the second full scale

test, as shown in Figure A. 53a. All loadcells were re - instrumented and

calibrated both with and without simulated surface roughness in the loading

head of the calibration test machine. The new cells exhibited linear output

(see Figure A. 54) and were relatively insensitive to surface irregularities.

Performance of these cells during the second full-scale test was substantially

improved and statics (sum of vertical forces) was met with generally

acceptable tolerances. Unfortunately, no data was available beyond 2 Ay

owing to failure of a circuit breaker supplying power to some 32 channels,

most of which were associated with the vertical load flatcells. Column

moment, as for the first test, was therefore determined on the basis of topof -column reactions. It was possible, however, using the available good

flatcell data to make some interesting observations regarding friction force,

since average friction force could also be calculated by knowing the reaction

frame force, the lateral ram force, and the load resisted by bending of the

test machine. A discussion of this topic is presented in Chapter 4.

Too often discussion of ideas which do not work is avoided in scientific

reports. The above difficulties are described with openess for the benefit of

those researchers who may need to monitor reactions where commercially

available hardware is not suitable. Suggestions for improved methods of

determining reactions are discussed in Appendix B.

A. 6.Tie -Back Frame

As a backup to the flatcells, the tie-back "A-Frame'' shown in Figure A. 23 was

instrumented with four full-bridge load cells. A typical calibration curve for

one of the four cells is shown in Figure A. 55. From statics, the moment at the

base of the column could be determined by knowing the A-Frame load, the

lateral displacement and applied axial load (for determination of p-delta

moment), and the lateral load dissipated into the test machine. During both

full scale tests, lateral displacement of the vertical loading head was

monitored. This displacement, in conjunction with specific bending stiffness

calibrations conducted for the vertical test machine in the .test configuration

for each specimen, allowed for calculation of lateral load resisted by the

226
18

061

227

Figure

A.

53

Typical

Calibration

for

Revised

Flatcell

Instrumentation

used

for

Full-Scale

Shear

Column

Test
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03

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228
test machine,

moment in lieu

The above approach was ultimately used to determine column

of direct friction measurements.

APPENDIX B

Recommendations for Future Full-Scale Test Apparatus and Procedures

As stated at the beginning of this report, conclusions were drawn from

comparisons between single prototpye specimens and models. Accurate

determination of the behavior of reinforced concrete elements from

experimental data requires a sufficient number of tests to establish

statistical bounds on the conclusions. For lifeline structures, such as

bridge columns, it would seem prudent to conduct further tests for

verification of the conclusions regarding similitude. The eventual payoff

will be the ability to use cost-effective models for prediction of prototype

behavior within specified limits of confidence. This would eliminate the need

for expensive verification (prototype) tests. The need for statistical

verification of the results presented in this report is particularly important

with regards to the conclusions concerning model construction materials (i.e.

that models constructed from small aggregate ready-mix concrete apparently

out-performed the microconcrete models)

.

When amortized over the entire cost of the program, including both labor and

hardware expenses, each prototype test cost approximately $500,000 in 1986

dollars. From the experience gained from these tests, it is believed that

production testing of prototypes could take place at substantially lower

costs

.

Early in the test program it was recognized that substantial labor and

materials costs were tied up in the fabrication of the massive mobile

foundation structure for each specimen. During testing this base block

usually developed only minor cracking and failure was localized in the column

itself and to a depth of a few inches into the base block. Strain gage data

indicated that the depth of longitudinal column reinforcement yielding was

limited to within 0.8 D into the foundation. Thus, pullout of longitudinal

reinforcing bars was not a factor, given the embedment depths presently

required by CALTRANS specifications. The implication is that the state of

stress in the foundation can be neglected in the assessent of column

231
performance, provided full development of longitudinal reinforcement is

scale specimen.

Given the foundation block does not effect to the performance of the column,

that it constitutes 85 % of the total weight of each specimen, and that the

block must be demolished and disposed of at great cost at the conclusion of

each test, it would seem advantageous to either eliminate it through the use * of alternative reaction fixtures or reuse portions of it for subsequent tests.

Figure B.l and B.2 show one possible method of making active use of the

majority of the specimen weight. Here, only the column itself and a thin

slice of the base block corresponding to the width of the column, is cast for

each full scale specimen. The remainder of the foundation block is comprised

of two precast blocks with shear keys on the interior faces to which the

column will be match cast. In fact, since the geometry of this joint does

not change from specimen to specimen, actual match casting is not required.

Lightweight metal forms can be made to serve the same purpose, allowing fast

tracking of casting and testing operations.

If it is assumed that the moving specimen concept is to be retained, the match

cast base blocks and central column specimen may then be made to perform as a monolithic unit by means of lateral post- tensioning of the base block, as

shown in Figure B.l. Consider the state of stress in such a situation.

First, the post- tensioning force in the base block must be sufficient to

prevent tension at maximum fiber distance along the two match-cast joints.

Figure B.2 shows the forces at work. M is the maximum moment developed by

the test column prior to failure and P is the applied axial load. The base

block is "b" units wide by "h" units high. The governing equation then

becomes

insured. This permits a reassessment of construction techniques for the full

F Fee Me

+ = 0 (Bl)

A I I

F (1 + ecbh/I

6 PL + M

(B2)

bh 2

232
P AXIAL

Figure B.l Match-cast Specimen Concept

233
MOMENTS ACTING ON FOUNDATION BLOCK

Column

Moment

Axial Load

Moment

Prestress

Moment

FOUNDATION BLOCK: BENDING & AXIAL INDUCED STRESSES

j~ _ FeC Me

A I !

Due to Due to

prestress prestress

direct load eccentricity

effect

Due to

external

moment

M q

Figure B.2 Moments Acting on Foundation Block for

Match-Cast Specimen Using Roller Supports

234
F

PL + M

(B3)

6 _ ( ) h 4 2

[ 1 + 6e/h]

For axial loads of 3000 kip (13.4 MN) , the corresponding maximum column moment

will be approximately 12,000 ft-kip (16.3 MN-m) . Assuming the same dimensions

i of the prototype specimens for L,b,e, and h (23.4, 8.0, 4.625, and 9.25 ft.

[7. 1,2. 4, 1.41, and 2.28 m] respectively), the required lateral posttensioning force is 4125 kip (18.4 MN) . In terms of actual hardware, this

level of force could be applied by using 27 of the 1-3/8 inch (35 mm) Dywidag

threadbars similar to those used for hardware connections in the present test

program. Post- tensioning of the precast segments together would likely

require 3 days of labor, a small price to pay considering the vast reduction

in effort compared to the present method.

Recall that the moving specimen concept was largely developed as a means of

utilizing an existing test apparatus (namely the NBS 12000 kip vertical test

machine) to the fullest possible extend in order to reduce program costs. The

above calculations indicate, clearly, that the moving specimen concept does

not lend itself easily to the notion of mass production testing. If one

started with a clean slate and could completely dissregard the particular

attributes of existing structural test facilities, what would be needed to

permit rapid, cost effective testing of bridge columns similar to those

described in this report?

First, consider the mechanics of the three part foundation block again. The

majority of the lateral post- tens ioning requirement, as defined in the above

calculations, results from the need to resist the large moments and shear

forces set up between the roller supports caused by the column axial load. If

the specimen rested against a reaction surface (as the 1/6-scale model

specimens did in the present study) then lateral post - tens ioning would only

need to counteract bending stresses set up by the column base moment

235
LEGEND: J

C

D

(Kxkxintr

T( Tiedown)

A - Upper hinge

B - Central hinge

C - Reaction structure for

axial load (upper)

D - Reaction structure for

axial load (lower)

E - Axial load ram

F - Lateral load ram

G - Column specimen

H - Match cast base blocks

J - Reaction structure for

lateral load

K - Reaction structure for

vertical load (lower)

—

/

K

Figure B.3 Alternative Servo-Mechanical Method for

Maintaining Axial Load in Vertical Plane

236
during testing. Surprisingly, this situation does not greatly reduce the

number of post- tensioning rod required. The reason is that without the large

flexural compression of the upper half of the block (previously provided by

the axial load) eccentric prestressing techniques cannot be used. The Fec/I

term in equation B1 vanishes, as does the PL/4 term, and M must be considered

for both clockwise and counterclockwise directions. This requirement calls

for stressing of 23 bars, a minor improvement over the previously derived 27

bars. Note that in this case the base block would also have to be transversely

post- tensioned to the reaction surface (see Figure B.3) to prevent sliding of

the specimen under the influence of lateral test loads. Transverse posttensioning would also be needed to prevent foundation uplift in the event that

the specimen is not tested in the upright position. Further discussion on

this latter option is presented below.

As described earlier in this report, the idea of applying all column forces to

the top of the column has substantial merit, and was the method of load

application for the model tests in which a computer-controlled loading head

was employed. If all column loads can be applied through a coincident hinge

at the top of the column, many sources for error are eliminated and direct

readings may be made of lateral and vertical loads by means of in-line axial

load cells. Friction effects are also eliminated. Certain important

considerations still exist, however.

Px.ecall that for realistic simulation of column axial (gravity) load during an

earthquake it is necessary to constantly maintain the load vector parallel to

the vertical plane. A "follower" load, such as might be achieved through the

use of a central axial post- tensioning tendon extending through the middle of

the column, does not accurately reproduce this loading state. Nor would the

use of a simple hinged actuator operating under load control without a special

geometric feedback feature in the control loop. These three cases are shown

graphically in Figure B.4. The above distinctions are drawn because in-situ

axial load for a bridge column is due to the dead weight of the

superstructure. The load path is always vertical and can give rise to

significant P-A effects which serve to accelerate failure under seismic load.

237
CL

05

CC

O

CL S

ID 05 § I ® TO

i: >> •4—5 o TO so S °

C/5

05 05 > o §

Q.

O CC — x 0 05

05

~o

05O

CDC

05

— •—

05 03 05 o

'$ I

238

Figure

B.4

Kinematics

of

Various

Methods

of

Applying

"Axial"

toa

Column
To ignore this fact during experimental testing would lead to liberal

estimations of energy absorption capacity.

For full-scale tests, two variations of a simple arrangement, such as shown in

Figures B.3 and B.5 offer methods for constructing a traveling load head. In

the first case, three servo -hydraulic actuators are used along with two

special hinges which permit coincident application of lateral and axial loads

to the column head. The two lateral actuators are driven by the same

displacement control signal, i.e. they both extend and retract the same

distance in concert during the course of a test. The bottom actuator (E) must

provide the lateral force necessary to fail the column; the top actuator (D)

need only overcome the friction force set up by the roller support at the top

hinge (A). The axial load actuator (which could, in fact, be a simple single

acting compression jack operated under variable drive pressure control [load

control requires a double acting jack]) is connected to both the top-ofcolumn hinge as well as to the top hinge (A) which bears against a reaction

surface with a roller support system (which can in all respects be similar to

the roller transport systems described in this report) . The upper lateral

actuator moves the top of the axial load actuator the same amount as the

lateral load ram (E) such that axial load actuator always remains in the

vertical plane.

This method requires the use of some sophisticated joints which would have to

be custom fabricated (and therefore costly) as well as three expensive

actuators. Furthermore, the upper reaction surface represents a substantial

engineering problem since a stiff overhead frame would have to be constructed.

On the positive side, no sophisticated control system is necessary to conduct

these tests: a simple, commonly available servocontroller can be used to

maintain load control on the axial load ram while identical displacement

control signals can be sent to the two lateral actuators. No geometry

corrections are necessary in the interpretation of the results and the

loadcell readouts represent true values for axial and lateral load, as well as

direct computation of column moment.

239
«

A physically simpler system is presented in Figure B.5. Here only two servocontrolled double-acting actuators are employed, each permanently attached to

a reaction surface and hinged at both ends. This is precisely the situation

mentioned above in which special consideration of the load system geometry

must be accounted for before sending a control signal to an actuator. The

situation is shown in Figure B.4, case b, where it can be seen that as the top

of the column is displaced laterally, the finite angle 9 between the vertical

axis and the instantaneous line of action of the axial load actuator gives

rise to both vertical and horizontal loads. However, this can be compensated

by providing the following control signals to each actuator:

P

Pl(i) -

c o s ( 9 )

p

A 2

cos[ sin'l ( )]

L1 + A 1

A 2 (i) lateral ram displacement at load stage "i" (inch)

(B4)

where

Pj_(i) = axial ram force (kip) command signal at load stage "i" under load

control. While shown here as a force, it would actually have to

be a digitally synthesized voltage sent to the servo

controller via a DAC unit on the computer.

A^ = piston extension in axial ram at load stage "i" (inch). Again,

this would be sent to the servo controller as a digitally

synthesized voltage.

P = desired constant (load control) axial column load (kip)

Lp = hinge - to -hinge lenth of axial load actuator prior to commencement

of testing (inch)

Actual implementation of the above algorithm would involve breaking each

desired lateral displacement move into many smaller moves (say 50) and

iteratively converging at each step to determine the required command signal

for the axial load ram so that constant vertical load is maintained. A

computerized data acquisition system is required to monitor load and

displacement feedback for both rams, as well as two channels of digital- to

240
LEGEND:

J

G

A - Upper hinge

B - Central hinge

C - Reaction structure for

axial load (upper)

D - Reaction structure for

axial load (lower)

E - Axial load ram

F - Lateral load ram

G - Column specimen

H - Match cast base blocks

J - Reaction structure for

lateral load

T( Tiedown) T

D

Figure B.5 Simplified Test Apparatus Employing

Computer-controlled Actuators

241
-analog output to the servo - controllers . Note that the the lateral actuator

must be capable of resisting potentially the large lateral, load component

generated by the axial load ram during large lateral displacements

particularly as column resistance begins to drop. Note that in the

calculation of column moment there are three terms, as follows:

Mtotal P 2 hc + Pqsin(#) + P A 2 (B5)

Where

^total

p 2

Pi

hc

9

Maximum Column Moment (kip -ft)

Load Cell Reading from Lateral Actuator (kip)

Load Cell Reading from Vertical Actuator (kip)

Column Height from Base to Top-of -column Hinge (feet)

Rotation Angle of Vertical Actuator from the vertical

(degrees, from equation [4])

Lateral Actuator Displacement (feet)

plane

Considering the low cost of personal computer (PC) based data acquisition

systems, for which analog- to - digital converters (ADC), digital - to - analog

converters (DAC) , and signal conditioning electronics are all presently

available as plug-in cards, the cost savings of the above approach will be

principally determined by the difference between the cost of programming the

control software and the funds saved by the elimination of expensive servo

hydraulic apparatus and test fixtures and their associated maintenance costs.

It should be noted that the test apparatus shown in Figure B.5 could equally

well be constructed in plan view rather than vertically as implied above.

This would eliminate access problems and would permit the use of a faci lity

with extensive tie-down floor capacity. Overturning moments would be

minimized in reaction fixtures. Furthermore, the requirement for large

overhead spaces would be obviated. Since individual specimen components

would be substantially lighter, a more versatile transport system, based on

air bearings, would be possible in lieu of a rail system.

242
STEEL

STRESS

(ksi)

APPENDIX C

Stress Strain Plots for Model and Prototype Reinforcing Steel

STRESS-STRAIN PLOT

FOR THE LONGITUDINAL REINFORCEMENT

Full-Scale Columns

Figure C.l: Longitudinal Reinforcement. Fnll- C c^le •- LONGITUDINAL STEEL FOR MODEL COLUMNS

0 0.02 0.04 0.06 0.08 0.10

STEEL STRAIN

Figure C.2: Longitudinal Reinforcement, Model Columns

243
STEEL

STRESS

(ksi)

SPIRAL REINFORCEMENT FOR FULL-SCALE FLEXURE COLUMN

Figure C.3: Spiral Reinforcement, Full-Scale Flexure Columns

SPIRAL REINFORCEMENT FOR FLEXURE MODEL COLUMN

(from straightened spiral specimens)

Figure C.4: Spiral Reinforcement, Model Flexure Columns

244
STEEL

STRESS

(ksi)

SPIRAL REINFORCEMENT FOR FULL-SCALE SHEAR COLUMNS

(from straightened spiral specimens)

P 2’_1 T

- e G 5’ ^ O "L 7" a 1 ^ sin ?OTrCSTr’ent TN 1 ^ 1 - do — G 0 G:-

'. —

SPIRAL REINFORCEMENT FOR SHEAR MODEL COLUMNS

from straightened spiral specimens)

rigure C.o: Spiral Reinforcement, Model Shear Columns

245

APPENDIX D

List of Symbols

Ac Area of column core, out-to-out of spiral (in^)

A

g

dh =

O

Gross area of column (in^)

Integration parameter for moment due to column weight,

where h is the height above the column base.

D = Diameter of column (in.)

D c “ Outside diameter of spiral (in.)

Ec = Modulus of elasticity of Concrete (ksi)

E s " Young's modulus of reinforcing steel (ksi)

Emc = Cyclic strain energy for model column (kip -inch)

[individual cycles]

Emt = Total cyclic strain energy for model columns (kip- inch)

[summation of all cycles until failure]

Epc = Cyclic strain energy for prototype column (kip- inch)

[individual cycles]

Ept = Total cyclic strain energy for prototype columns (kip- inch)

[summation of all cycles until failure]

fL c = Concrete compressive strength (ksi)

f' cc = Maximum confined strength of concrete (ksi)

^em = Energy modification factor

- fie 3

^le — Load effectiveness ratio

= Ee ff / Fram

^su = Ultimate strength of longitudinal reinforcing steel (ksi)

f

y

=

f l -

Yield stress of steel (ksi)

Effective lateral confining pressure on concrete (ksi)

Fe ££ = Actual lateral load applied to the column (kip)

Fj_ = Friction force at i(th) roller support (kip)

Fram

= Lateral load applied to the column base block by the hydraulic

h0 -

actuator (kip)

Height of specimen base block (in.)

h l = Effective height of test column (in.)

247
h-2 = Height of top-of -column hinge axis (in.)

h3 = h2 - hp ( in.

h^ = Vertical test machine crosshead height (in.)

Km

= Vertical test machine bending

stiffness (kip/in.)

1 ^ = Development length of rebar (in.)

Lp = Plastic hinge length (in.)

Lp = Lateral offset for tie-back frame hinge (in.)

L9 = Center- to-center distance between

roller supports (in.)

M = Column base moment (kip -in.)

Mum = ultimate moment for model column (kip -inch)

M^ = ultimate moment for prototype column (kip- inch)

M]_ = Column base moment as calculated

from top of column reactions (kip-in.)

M2 = Column base moment as calculated

from foundation reactions (kip-in.)

= Vertical reaction at i(th) roller support (kip)

P a = Axial load (kip)

P a f= Reaction wall load (kip)

P e = Axial load on column (kip)

Pm = Load dissipated into vertical test machine (kip)

P r = Lateral actuator load (kip)

s ld = experimental scale factor for yield deflection

Ayp/A

ym

s le = experimental scale factor for absorbed strain energy

- (Epc/Emc ) 1/3

" (Epc/Emt ) 1/3

s lm ^ experimental scale factor for ultimate moment

= ^^up/^um^^^

V = Column base shear (kip)

V]_ = Column base shear as calculated

from top of column reactions (kip)

y 2 = Column base shear as calculated

from top of column reactions (kip)

248
W = Specimen weight (including base block) (kip) = Unit weight of column (kip/in.)

a ^provided / ^code

= Vertical offset of tie-back hinge

due to top-of -column rotation (in.)

A(h) = Lateral deflection of column @ height "h" (in.)

Ar = Lateral deflection at base of column (in.)

A t = Total displacement for computation of

P-A moment ( in.

)

Ay = Yield deflection (in.)

Ayuj = Yield deflection for model column (in.)

Ayp = Yield deflection for prototype column (in.)

6 sh = Steel strain at the onset of strain hardening

e su

= Maximum steel strain

= Tensile strain as measured by clip gage at height i

e 2i = Compressive strain as measured by clip gage at height i

H = Displacement ductility

fj.^_ = Coefficient of friction at i(th) roller support

4> = Strength reduction factor for confined members

4>y = Yield curvature (rad/in.)

p C ode = Volumetric transverse steel ratio required by NZ-3101 [20]

^provided = Volumetric transverse steel ratio provided for NBS

specimens as guided by standard CALTANS practice

p s = Volumetric or transverse steel ratio

p t = Reinforcing ratio for longitudinal steel

test

249
* ' 1, . ,
REFERENCES

1. Cheok, Geraldine S., and Stone, William C., "Behavior of 1/6 - Scale Model

Bridge Columns Subjected to Cyclic Inelastic Loading", NBSIR 86-3494, U.S.

Department of Commerce, November, 1986.

2. "Recommendations for Future Research", Applied Technology Council Workshop

on Earthquake Resistance of Highway Bridges, Jan. 1979. Applied

Technology Council, Berkeley, California. 625pp.

3. Lew, H. S., Leyendecker, E. V., and Dikker, R. D ., "Engineering Aspects of

the 1971 San Fernando Earthquake, NBS Building Science Series #40, U.S.

Department of Commerce, Dec. 1971.

i 4. "Standard Specifications for Highway Bridges", as amended by AASTHO

interims through 1984 with revisions by CALTRANS , AASHTO , 1983.

5. "Seismic Design Guideline for Highway Bridges", Applied Technology

Council, Berkeley, California, October 1981. 200pp.

6. "Exerpts from the Standard Specifications for Highway Bridges Relating

to Seismic Design," 13th Edition, 1983 Adopted by the American

Association of State Highway and Transportation Officials with

Revisions by CALTRANS, Division of Structures. Version October 1985.

7. Ang, Beng Ghee, "Ductility of Reinforced Concrete Bridge Piers Under

Seismic Loading," Research Report 81-3, February 1981, Department of

Civil Engineering, University of Canterbury, Christchurch, New Zealand

8. ACI Structures Journal Nov. 1968, Feb 1971

9. Montgomery, Neil, "Single Rope Techniques," Available from the

National Speleological Society Bookstore, 1 Cave Ave . , Hunstville,

Al. 1977.

10. Padget, Alan, et.al., "On Rope," Published by the National

Speleological Society, 1 Cave Ave., Huntsville, Al . 1987.

11. ACI Committee 340, "Design Handbook," in accordance with the Strength

Design Method of ACI 318-77, Vol. 2, SP-17A (78), American Concrete

Institute, 1978.

12. Personal Communication with Dr. M.J.N. Priestly, University of

Canterbury, 1985.

13. "Recommended Testing Procedure for Assessing the Behavior of Structural

Steel Elements under Cyclic Loads," ECCS -Technical Committee 1,

Structural Safety and Loadings Technical Working Group 1.3 - Seismic

Design Report. First Edition No. 45, 1986. European Convention for

Constructional Steelwork. Available from A. Plumier, Institut du Genie

Civil, 6 Quai Banning, Universite de Liege, Belgium.

251
I

I

14. Krawinkler, H. et.al., "Recommendations for Experimental Studies on the

Seismic Behavior of Steel Components and Materials," Report No. 61, The

John A. Blume Earthquake Engineering Center, Department of Civil

Engineering, Stanford University, September 1983.

15. Mander, J.B., "Seismic Design of Bridge Piers," Research Report 84-2.

February 1984, Department of Civil Engineering, University of Canterbury,

Christchurch, New Zealand.

16. Priestly, M.J.N., and Park, R. , "Strength and Ductility of Bridge

Substructures," Research Report 84-20, December 1984. Department of

Civil Engineering, University of Canterbury, Christchurch, New Zealand

17. Priestly, M.J.N., and Park, R. , "Strength and Ductility of Concrete

Bridge Columns Under Seismic Loading," ACI Structural Journal, Title

No. 84-S8, Vol. 84, No. 1, January- February 1987. American Concrete

Institute, Detroit, MI.

18. Personal communication, Mr. James Gates, Chief Engineer, Structures

Division, CALTRANS , 1983.

19. Stone, W.C., and Breen, J.E., "Behavior of Post-Tensioned Girder

Anchorage Zones," PCI Journal, Jan- Feb, 1984, Vol. 29, No. 1,

pp 64-109. Prestressed Concrete Institute, Chicago, 111.

20. Baker, A.L., and Amarakone , A.M., "Inelastic Hyperstatic Frames

Analysis," Proceedings of the International Symposium on the

Flexural Mechanics of Reinforced Concrete, ASCE-ACI, Miami,

November 1964, pp 85-142.

21. Corley, W.G., "Rotational Capacity of Reinforced Concrete Beams,"

Journal of Structural Division, ASCE, Vol. 92, STS, October 1966,

pp 121-146.

22. "NZS-3101: 1982 'Code of Practice for the Design of Concrete

Structures'", (Two Volumes: Code and Commentary), Standards

Association of New Zealand, 1982.

252
N B S - 1 14A rev. 2 -ac) U.S. OEPT. OF COMM.

BIBLIOGRAPHIC DATA

SHEET (See instruction s)

1. PUBLICATION OR

REPORT NO.

NBSIR 88-3788

2. Performing Organ. Report No 3. Publication Date

MAY 1988

4.

TITLE AND SUBTITLE

INELASTIC BEHAVIOR OF FULL-SCALE BRIDGE COLUMNS SUBJECTED TO CYCLIC LOADING

5.

AUTHOR(S)

WILLIAM C. STONE and GERALDINE S. CHEOK

6.

PERFORMING ORGANIZATION (If joint or other thanNBS, see instructions)

NATIONAL BUREAU OF STANDARDS

U.S. DEPARTMENT OF COMMERCE

GAITHERSBURG, MD 20899

7. Contract/Grant No.

3. Type of Report & Period Covered

9.

SPONSORING ORGANIZATION NAME AND COMPLETE ADDRESS (Street. City , State, ZIP)

1. National Science Foundation (NSF)

2. Federal Highway Administration (FHWA)

3. California Department of Transportation (CALTRANS)

4. National Bureau of Standards (NBS)

10.

SUPPLEMENTARY NOTES

Document describes a computer program; SF-185, FlPS Software Summary, is attached.

11.

ABSTRACT (A 200-word or iess factual summary of most si gn i fi cant information. If document includes a significant

bi bliograohy or literature survey, mention it here/

Circular , spirally reinforced concrete bridge columns were subjected to cyclic

loading (representing seismic loads) in the laboratory. The test articles were

prototype columns designed in accordance with recent (1983) California Department

of Transportation (CALTRANS) specifications. Two full-scale columns, each

measuring 5 feet (1.52m) in diameter with aspect ratios (height 'width) of 3 and 6,

were subjected to slow reversed cyclic lateral load with constant axial load to simulate the gravity weight of the bridge superstructure. Details are presented concerning the design of the special test apparatus required to conduct the project as

well as recommendations for future improvements in test orocedures.

Results from the tests are presented in the form of energy absorption granhs and bar

charts, load-displacement hysteresis curves, longitudinal and confining steel strains,

and curvature profiles. Comparisons of the ultimate moment capacities, measured

displacement ductilities, plastic hinge lengths, cyclic energy absorption capacity

and the failure modes of the full-scale specimens are made with those observed from

1/6-scale model tests.

12.

KEY WORDS (Six to twelve entries; alphabetical order; capitalize only proper names; and separate key words by semicolons/

Axial Load; Behavior; Bridges; Columns; Concrete; Confinement; Cyclic Load; Ductility;

Energy Absorption; Failure; Full-Scale; Lateral Load; Microconcrete; Scale Effects

14. NO. OF

PRINTED PAGES

262

15. Price

S24.95

13.

AVAILABILITY

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