The Lorna Prieta, California, Earthquake of 

October 17, 1989-Earth Structures and 

Engineering Characterization of Ground 

Motion 

THOMAS L. HOLZER, Editor 

PERFORMANCE OF THE BUILT ENVIRONMENT 

THOMAS L. HOLZER, Coordinator 

UeSe GEOLOGICAL SURVEY PROFESSIONAL PAPER 1552-D 

Prepared in cooperation with the National Science Foundation 

UNITED STATES GOVERNMENT PRINTING OFFICE, WASHINGTON : 1998 

CONTENTS 

Performance of earth dams during the Lorna Prieta earthquake ----------- 3 

By L.F. Harder, J.D. Bray, R.L. Volpe, and K.V. Rodda 

Analysis of soil-nailed excavations stability during the 1989 Lorna 

Prieta earthquake ............................................................... 27 

By Mladen Vucetic, Mark R. Tufenkjian, Guy Y. Felio, Pirooz Barar and 

K. Ronald Chapman 

Empirical analysis of peak horizontal acceleration, peak horizontal 

Attenuation of vertical and horizontal response spectra of the Lorna 

Prieta earthquake ............................................................... 69 

By Yousef Bozorgnia and Mansour Niazi 
THE LOMA PRIETA, CALIFORNIA, EARTHQUAKE OF OCTOBER 17,1989: 

PERFORMANCE OF THE BUILT ENVIRONMENT 

EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

INTRODUCTION 

By Thomas L. Holzer, 

U.S. Geological Survey 

This chapter contains two papers that summarize the performance of engineered earth structures-dams and stabilized excavations in soil-and two papers that characterize 

for engineering purposes the attenuation of ground motion 

with distance during the Loma Prieta earthquake. Documenting the field performance of engineered structures and confirming empirically based predictions of ground motion are 

critical for safe and cost effective seismic design of future 

structures as well as the retrofitting of existing ones. 

Catastrophic flooding caused by dam failure is a persistent concern in areas with earthquake potential. Harder and 

others (this chapter) present a valuable compilation of the 

performance of earth dams during the 1989 earthquake. They 

estimate that 11 1 dams, the majority of which were homogenous earth dams, were within 80 km of the seismogenic 

rupture. Thirty-six of these dams experienced free-field peak 

horizontal ground accelerations greater than 0.2 g, but only 

one major dam-Austrian-and one minor dam-Soda 

Lake-suffered moderate damage. The authors conclude that 

low reservoir levels, which reduced saturation of embankments, was a major contributor to the good performance of 

these dams. The field performance and strong-motion data 

provide an opportunity to learn more about how earth and 

rockfill dams behave during earthquakes at low reservoir 

levels. 

Failures of embankments in soils during earthquakes both 

disrupt traffic and damage nearby buildings. Vucetic and 

others (this chapter) evaluate the performance of nine excavations that were stabilized by soil nailing, an in-situ technique used to reinforce existing soil masses during excavation by horizontally drilling and installing passive inclusions. 

The inclusions, called nails, cause the soil to behave as a 

composite unit. The 1989 earthquake provided the first seismic loading of earth structures of this type and thus presents 

a special opportunity to evaluate the performance of nailed 

walls. None of the nine soil-nailed excavations evaluated by 

the authors showed any signs of distress or movement even 

though one in the epicentral regional was strongly shaken. 

The authors conclude that conservative design and construction are the primary reasons for their good performance. 

The most common methods for estimating ground motion 

in future earthquakes in California are empirical and rely on 

syntheses of observations from previous earthquakes. Papers by Campbell (this chapter) and Bozorgnia and Niazi 

(this chapter) build on this tradition. For further discussion 

of other aspects of strong ground motion during the earthquake, the reader is referred to the 16 papers in Borcherdt 

(1994) as well as selected papers in Spudich (1996). 

Campbell (this chapter) examines the dependence of recorded peak horizontal acceleration, peak horizontal velocity, and Modified Mercalli Intensity on distance from the 

seismogenic rupture zone, source-to-site azimuth, and site 

geology. He develops empirical fits to these parameters. He 

concludes that peak accelerations recorded on alluvium at 

sites more than 50 km from the seismogenic rupture zone 

were significantly higher than would have been predicted 

by existing attenuation relations available at the time of the 

earthquake. He also notes a strong directional or azimuthal 

dependence of ground motion and intensity. Significantly 

higher amplitudes corresponded to azimuths pointed at Santa 

Cruz, San Francisco, and Oakland than at azimuths to the 

northeast and east. Local site geology strongly affected 

ground motion with sites underlain by Holocene San Francisco bay mud typically recording higher accelerations. 

Bozorgnia and Niazi (this chapter) study the frequency 

dependence of the attenuation of ground motion by examining vertical and horizontal response spectra for spectral ordinates at periods ranging from 0.05 to 2 s. They observe 

that the shape of the response spectra for both vertical and 

horizontal components of ground motion and their ratio depends on both distance and magnitude. The vertical-to-horizontal spectral ratio at higher frequencies is substantially 

higher than the two-thirds value commonly used in engineering practice. The ratio is highest near the earthquake 

source. However, at low frequencies and farther from the 

earthquake source, the spectral ratio decreases significantly, 

which implies current design practice is conservative. 

REFERENCES CITED 

Borcherdt, R.B., ed., 1994, The Lorna Prieta, California, earthquake 

of October 17, 1989-Strong ground motion: U.S. Geological Professional Paper 155 1-A, 272 p. 

Spudich, P.A., ed., 1996, The Lorna Prieta, California, earthquake 

of October 17,1989-Main-shock characteristics: U.S. Geological Professional Paper 1550-A, 297 p. 
THE LOMA PRIETA, CALIFORNIA, EARTHQUAKE OF OCTOBER 17,1989: 

PERFORMANCE OF THE BUILT ENVIRONMENT 

EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

PERFORMANCE OF EARTH DAMS DURING THE 

LOMA PRIETA EARTHQUAKE 

By L.F. Harder, Jr., California Department of Water Resources, Sacramento, Calif.; 

J.D. Bray, University of California, Berkeley, Calif.; 

R.L. Volpe, R.L. Volpe & Associates, Los Gatos, Calif.; and 

K.V. Rodda, Wahler Associates, Palo Alto, Calif. 

CONTENTS 

Page 

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6 

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8 

13 

13 

15 

15 

17 

17 

18 

18 

21 

2 1 

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21 

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23 

24 

25 

25 

25 

ABSTRACT 

The earthquake shook a large number of earth and rockfill 

dams. There were more than 100 dams within 50 miles of 

the fault rupture associated with this event. Although more 

than half of the dams were less than 60 feet in height, a number of major dams were strongly shaken. In general, the dams 

performed satisfactory with one major dam and one minor 

dam developing moderate damage. A small number also developed minor to moderate cracking and required repairs. 

The great majority, however, sustained no significant damage. Although this result was very encouraging, it should be 

tempered by the fact that the reservoirs retained by many of 

these dams were low at the time of the earthquake. In addition, the duration of strong shaking during this earthquake 

was relatively short. Consequently, the earthquake was not 

the most critical test of these structures. 

INTRODUCTION 

The earthquake resulted from a slip on a segment of the 

San Andreas fault near Santa Cruz, Calif. As summarized 

by Spudich (1996), slip occurred on a dipping fault surface 

approximately 22 miles long and was generally confined to 

depths between 4 and 12 miles. From the hypocenter, slip 

propagated in both the northwest and southeast directions 

over a relatively short source duration of approximately 6 

to15 s. Most studies suggest that slip in the southeast portion 

was predominantly strike-slip, whereas slip in the northwest 

portion also had a significant thrust component. 

Most of the slip associated with the main shock is considered to have stopped within 4 miles of the ground surface, 

although surface cracking observed in the area near the fault 

may have been associated with zones of compression and 

extension generated by a tectonically offset basement 

(Spudich, 1996). The event represents the largest earthquake 

in the San FranciscoISanta Cruz area since the great 1906 

San Francisco earthquake of magnitude 8+. The Loma Prieta 

fault rupture occurred along a portion of the San Andreas 

fault segment which ruptured during the 1906 earthquake. 

The duration of strong shaking for this earthquake was 

generally between 7 and 10 s, about half of that usually associated with a magnitude 7 event. It has been speculated 

that this short duration was a result of the central location of 

the earthquake's focal point and its bidirectional rupture pattern (see Seed and others, 1990). The highest horizontal 

ground acceleration recorded was 0.64 g and was measured 

at the Corralitos station located adjacent to the surface projection of the fault rupture and only a few miles from the 

epicenter. Seismographs in the epicentral area also recorded 

relatively high vertical accelerations that were comparable 

to those recorded in the horizontal direction. For example, at 
D4 PERFORMANCE OF THE BUILT ENVIRONMENT 

the Capitola recording station, located approximately 9 miles 

southwest of the epicenter, the peak vertical acceleration was 

0.60 g whereas the peak horizontal acceleration was 0.54 g. 

CHARACTERISTICS OF EARTH DAMS 

CONSIDERED IN THIS STUDY 

Presented in figure 1 is a map showing the epicenter and 

locations of major aftershocks associated with the earthquake. 

Also shown in this figure are the locations of 11 1 earth dams 

situated within 50 miles of the fault rupture zone. The majority of these dams are essentially homogeneous earth dams. 

The heights and completion dates for these dams are summarized in tables 1 and 2. 

Table 3 presents a listing of the number of dams subjected 

to various levels of ground shaking during the earthquake. 

The estimates of peak ground acceleration were developed 

by interpolating between values measured at nearby seismographic stations. 

Most of the reservoirs in the affected area were relatively 

low at the time of the earthquake, with many at less than half 

their normal heights. Some reservoirs were essentially empty. 

This was a result of the fact that the earthquake occurred 

immediately after the irrigation season following 3 years of 

lower-than-average rainfall. 

PRE-1906 EARTH DAMS 

Of the 21 dams built prior to the 1906 earthquake, none 

experienced significant damage as a result of the Loma Prieta 

earthquake. This is not unexpected, as these dams generally 

performed satisfactorily during the 1906 earthquake. Because 

the 1906 earthquake involved a much greater fault rupture 

and release of energy, all of these older dams had probably 

been subjected to much greater amplitudes and durations of 

shaking during that event than during Loma Prieta. Many of 

these older dams were located relatively close to the 1906 

fault rupture and have been estimated to have been shaken 

by motions having peak accelerations between 0.6 and 0.8 g 

(Seed and others, 1978). 

In their examination of the performance of earth dams 

during earthquakes, Seed and others (1978) attributed the 

good performance of these older dams during the 1906 earthquake to the fact that they were constructed mostly of clayey 

soils and built on either clayey or rock foundations. Such 

materials are generally thought to retain most of their 

strengths during strong shaking. 

Two of these older dams, University Mound N. Basin and 

Piedmont Dams, were built primarily of sandy soils but were 

not thought to have been saturated at the time of the 1906 earthquake. These two dams are located more than 45 miles from 

the Loma Prieta fault-rupture zone and experienced peak ground 

accelerations of only about 0.1 g with no reported damage. 

Table 1 .-Maximum heights of earth dams considered in this study 

Maximum height (ft.) Number of dams 

Total 111dams 

Table 2.-Dates of completion for earth dams considered in this study 

Completion dates Number of dams 

Total 11 1 dams 

Table 3.-Estimates of peak ground accelerations for earth dams 

considered in this study 

Estimated peak ground Number of dams 

acceleration (g) 

Total 11 1 dams 
PERFORMANCE OF EARTH DAMS DURING THE LOMA PRIETA EARTHQUAKE D5 

;AN 

FRAN 

.. 

. . .:. .: \ 

\ 

HAY WARD RMORE 

\ 

\ 

RCALAVERAS DAM s^ 

<>&"ILL :. -:: DAM 

: SANLUISDAM \ .... ... 1 37O00' 

I 

ALINAS V ^ 

"^- Â A. 

9':. 

<>. 

/ . . / 

/ 

/ . / 36' 3C 

Figure 1 .-Locations of 11 1 earth and rockfill dams within 50 miles of the Loma Prieta fault rupture zone (diagonal line pattern). 
PERFORMANCE OF THE BUILT ENVIRONMENT 

HYDRAULIC FILL DAMS 

There are five hydraulic fill dams within 50 miles of the 

fault rupture associated with Loma Prieta: Mill Creek, 

Chabot, Calaveras, Hawkins, and Old Upper San Leandro. 

Chabot Dam is one of the predominantly clayey dams 

which performed well during the 1906 earthquake. This dam 

was located about 37 miles from the Loma Prieta fault-rupture zone and experienced a peak ground acceleration of about 

0.1 g without sustaining significant damage. 

The Old Upper San Leandro Dam was replaced in 1977 

with a new dam immediately downstream of the previous one. 

Both structures were located about 41 miles from the Loma 

Prieta fault-rupture zone and experienced peak ground accelerations of about 0.1 g without sustaining significant damage. 

The performance of the Calaveras, Mill Creek, and 

Hawkins Dams is discussed in subsequent sections. 

ROCKFILL DAMS 

No dams are composed completely of rockfill within 50 

miles of the Loma Prieta fault- rupture zone. However, many 

earth dams in this area were built with substantial zones of 

rockfill. Even Calaveras Dam, known as a hydraulic fill, was 

completed with substantial zones of rockfill. The performance 

Table 4.-Hydraulic fill dams located within 50 miles of the Lorna 

Prieta fault rupture zone 

Dam Year Max. height 

completed (feet) 

Mill Creek 1889 76 

Chabot 1892 142 

Calaveras 1925 210 

Hawkins 1928 72 

Old Upper San Leandro - - -- 

of some prominent dams built with zones of rockfill is discussed in later sections. 

SELECTED CASE HISTORIES OF 

PERFORMANCE 

Information regarding the performance of 35 earth dams 

located relatively close to the fault rupture is presented in 

table 5. The locations of these dams are shown in figure 2. 

Additional details for selected dams are presented in following sections. 

LAUREL SPRING 

CLUB 

Â 

MURRY 

,CO I T 

ELMER J Â I KELLY CABIN 

CHESBRO CoYOT~ CANYON 

Â¥R SIMON1 IRR 

Â UVAS 

PEABODY Â 

SELVAGE NO. 2 NORTH 

FORK 

0 

HAWKINS 

Â¥VESSE 

. HOLLISTERÃ 

* 

Â 

^ 

HOLLISTER 

*SAN JUST0 . . 

Figure 2.-Locations of earth and rockfill dams in close proximity to the Loma Prieta fault rupture zone. 
PERFORMANCE OF EARTH DAMS DURING THE LOMA PRIETA EARTHQUAKE 

,' 

Table 5.-Summary of performance for selected case histories during the 1989 Loma Prieta earthquake 

DAM 1 1 DATE OF 1 MAX. HEIGHT 1 APPROX. SOURCE I ESTIMATED 1 DAMAGE 1 1 COMPLETION 1 (feet) 1 DISTANCE (miles) 1 PGA (g) 1 

AUSTRIAN 1 ERTH I 1950 1 185 1 < 1 1 0.55 - 0.6 1 MODERATE SETTLEMENT, TRANSVERSE AND LONGITUDINAL CRACKING, SPILLWAY DAMAGE 

-- 

COWâ L L ERTHI 1890 1 50 1 11. I 0.45 - 0.5 I NO REPORTED DAMAGE, RESERVOIR EMPTY 

BAY STREET 1 ERTH 1 1924 1 27 1 10. 1 0.45 - 0.5 1 MINOR SURFICIAL CRACKING ON CREST 

DEBELL 1 ERTH 1 1952 1 53 1 4.5 10.45-0.5 INOREPORTEDDAMAGE 

LEXINGTON 1 ERTH 1 1953 1 205 1 2. 1 0.45 1 MODERATE LONGITUDINAL AND TRANSVERSE CRACKING, 0.9 FEET OF SETTLEMENT 

- - 

LWER HOWELL I ERTH I 1877 1 39 1 1.5 1 0.4 - 0.45 1 1/4 - INCH LONGITUDINAL CRACK ON CREST 

UPPER HOWELL 1 ERTH I 1878 I 36 1 1.5 10.4 -0.45lNOREPORTEDDAMAGE 

ALMADEN 1 ERTH 1 1936 1 110 1 6. 1 0.4 - 0.45 1 MINOR LONGITUDINAL CRACK AT JUNCTION OF DAM CREST AND U/S CONCRETE FACING 

UVAS 1 ERTH 1 1957 1 118 1 5.5 1 0.4 - 0.45 I NO REPORTED DAMAGE 

-- 

R SIMON1 IRRIGATION 1 ERTH 1 1961 1 44 1 6. I 0.4 - 0.45 1 NO REPORTED DAMAGE, RESERVOIR VIRTUALLY EMPTY 

GUADALUPE 1 ERTH 1 1935 1 142 1 6. 1 0.4 - 0.45 1 MODERATE CRACKING AT TOP OF U/S BERM, SPALLING OF U/S CONCRETE PANELS 

LAKE RANCH 1 ERTH 1 1877 1 38 1 4. 1 0.4 - 0.45 1 2 EMBANKMENTS WITH MINOR LONGITUDINAL AND TRANSVERSE CRACKING 

NEWELL 1 ERRK 1 1960 1 182 1 6. 1 0.4 - 0.45 1 I TO 9- INCH WIDE LONGITUDINAL CRACKS IN UPPER U/S SLOPE, INCREASED SEEPAGE 

- -- - - 

ALMADEN VALLEY I ERTH 1 1965 1 38 1 7. 1 0.4 - 0.45 1 NO REPORTED DAMAGE 

ELMER J. CHESBRO 1 ERTH 1 1955 1 95 1 8. 1 0.4 - 0.45 1 MODERATE LONGITUDINAL CREST CRACKING, MINOR TRANSVERSE CRACKING AT ABUTMENT 

CALERO 1 ERTH 1 1935 1 90 1 8. 1 0.35 - 0.4 1 NO REPORTED DAMAGE 

VASONA PERCOLATION 1 ERTH 1 1935 1 34 1 5.5 I 0.35 - 0.4 1 MINOR LONGITUDINAL CREST CRACKING, MINOR TRANSVERSE CRACKING NEAR SPILLWAY 

SELVAGE NO. 2 1 ERTH 1 1948 1 42 1 7. 1 0.35 - 0.4 1 NO REPORTED DAMAGE 

R I NCONADA 1 ERTH I 1969 1 40 1 6. 1 0.35 - 0.4 1 4 TEARS IN RUBBER LINER SEAMS, CRACKED CONCRETE INTAKE VAULT 

VESSEY 1 ERTH 1 1945 1 20 1 13. 1 0.3 - 0.35 1 MINOR LONGITUDINAL CRACKING ON CREST, RESERVOIR EMPTY 

OAK SITE 1 ERTH 1 1969 1 43 1 9.5 10.3 -0.35lNOREPORTEDDAMAGE 

SODA LAKE 1 ERTH 1 1978 1 35 1 5.5 I 0.3 - 0.35 1 4 TAILINGS DAMS - LIQUEFACTION OF TAILINGS, SLUMPING OF 10-FT WEST DAM 

MILL CREEK 1 HYDF 1 1889 1 76 1 12. 1 0.25 - 0.3 1 MINOR LONGITUDINAL CRACKS ON CREST 

1 SEMPERVIRENS ERTHI 1951 1 42 1 12. 1 0.25 - 0.3 1 NO REPORTED DAMAGE I HOLLISTER UU PD I ERTH 1 1972 I 13 1 16. 1 0.25 - 0.3 1 MODERATE CRACKING, RESERVOIR EMPTY 

SAN JUST0 1 ERRK 1 1986 1 135 1 17. 1 0.26 1 NO REPORTED DAMAGE 

LEROY ANDERSON 1 ERRK 1 1950 1 235 1 13. 1 0.26 1 MINOR LONGITUDINAL CRACKING ON NEW CREST FILL 

HAWK INS 1 HYDF 1 1931 1 72 1 21. 1 0.2 - 0.25 1 MINOR LONGITUDINAL CRACKS ON CREST AND U/S SLOPE, RESERVOIR EMPTY 

I PEABODY ERTHI 1950 1 63 1 12. 1 0.2 - 0.25 1 NO REPORTED DAMAGE 

--- - 

COYOTE ERRKI 1936 1 140 1 14. 1 0.19 INOSIGNIFICANTDAMAGE 

KELLY CABIN CANYON 1 ERTH 1 1955 1 32 1 18. 1 0.15 - 0.2 1 NO REPORTED DAMAGE 1 COIT 1 ERTH 1 1956 1 1 19. 1 0.15 - 0.2 1 NO REPORTED DAMAGE 

LAUREL SPRING CLUB 1 ERTH 1 1968 1 28 1 20. 0.15 - 0.2 1 NO REPORTED DAMAGE 

1 MURRY 1 ERTH 1 1957 1 54 1 22. 1 0.1 - 0.15 I NO REPORTED DAMAGE 1 

1 NORTH FORK ERTHI 1939 1 100 1 23. 1 0.1 - 0.15 1 NO REPORTED DAMAGE, RESERVOIR VIRTUALLY EMPTY 

NOTES: ERTH = EARTH, ERRK = EARTH AND ROCKFILL, HYDF = HYDRAULIC FILL 
PERFORMANCE OF THE BUILT ENVIRONMENT 

AUSTRIAN DAM 

Austrian Dam was the earth dam most heavily damaged by 

the Loma Prieta earthquake. This dam is located directly above 

the projected northern segment of the fault rupture and about 7 

miles from the epicenter (fig. 2). Because of its proximity to 

the earthquake, it is also thought to have been the dam to have 

experienced the largest shaking with peak ground accelerations 

estimated to have been between 0.55 and 0.6 g. 

Fortunately, at the time of the earthquake, the reservoir 

elevation was about 100 feet below the dam crest. In fact, 

the reservoir level had been depressed during a 3 to 4 year 

drought prior to the earthquake, and the upper embankment 

materials were not fully saturated. The damage sustained by 

the embankment included moderate settlement, downstream 

movement, and moderate longitudinal and transverse cracking (fig. 3). The concrete spillway located on the right abutment was also heavily damaged. Details of the investigations and remedial construction are described in a report by 

Wahler Associates (1990). Previous summaries of damage 

were presented by Bureau and others (1989), Rodda and others (1990), and Seed and others (1990). 

Austrian Dam is a 185-foot-high rolled earth fill dam and 

was completed in 1950. The dam site is situated between 

the San Andreas and the Sargent faults. The dam is about 

4,000 feet southeast of the main intersection of these fault 

zones, with the trace of the 1906 movement on the San 

Andreas fault located only 1,700 feet south of the dam. The 

Sargent fault is located less than 700 feet north of the dam. 

The dam is founded on rocks typical of the Franciscan Complex, including highly fractured sandstone, graywacke, 

cobble conglomerate, shale, and serpentine. 

AREA OF LONGITUDINAL 

AREA OF TRANSVERSE AND OBLIQUE CRACKING 

. LOCATION OF PRE-EARTHQUAKE 

PIEZOMETERS 

o 50 I00 150 200feet SCALE - ' 

Figure 3.-Plan view of Austrian Dam locating cracks induced by the earthquake-section A-A'shown in figure 4 (adapted from Wahler 

Associates, 1990). 
PERFORMANCE OF EARTH DAMS DURING THE LOMA PRIETA EARTHQUAKE 

I V/u 

RESERVOIR ELEVATION 

ON OCT. 17, 1989 %I024 feet .5-. 

-- -7-, 

IMPERVIOUS ZONE 

WITH GROUT CURTAINS 

Figure 4.-Maximum cross section of Austrian Dam (section A-A' adapted from Wahler Associates, 1990). 

The embankment was built by selective borrowing in an 

attempt to create a more impervious upstream zone in comparison to the downstream half of the embankment. Gravel 

strip drains were also placed beneath the downstream "pervious" zone. Figure 4 presents a view of the maximum cross 

section, together with the approximate reservoir and piezometer levels on the day of the earthquake. 

Sampling of the embankment materials during earlier studies of the dam and the piezometer readings together indicate 

that there is not an appreciable difference between the upstream "impervious" and downstream "pervious" zones, and 

that the gravel drains are not completely effective in relieving downstream seepage pressures. Hence, the dam can be 

considered to be nearly homogeneous. 

Selected embankment material properties are summarized 

in table 6. This table presents gradation, compacted density, 

and placement moisture content values from control tests 

performed during the original construction in 1950, together 

with dry density results measured in the dam following the 

earthquake. Table 6 also presents plasticity, shear strength, 

and values from laboratory tests performed on undisturbed tube samples taken in 1979. During the remedial construction in 1989, the average relative compaction measured 

in the preexisting upstream shell of the dam (based on a 

compactive effort of 20,000 ft.-lb/ft3) was 93 percent. As 

shown in table 6, the in-situ dry densities of the fill materials 

in those tests were generally in the upper half of the range of 

dry densities determined during the 1950 construction control tests. 

Significant ground movement in the vicinity of the dam 

was observed on the downstream right abutment above the 

spillway chute, on the upstream left abutment between the 

dam crest and the inclined inlet, and along the right side of 

the reservoir upstream of the dam (fig. 3). 

On the right abutment, a nearly continuous arcuate scarp 

was observed, extending up to about 200 feet above the dam 

crest and about 1,000 feet downstream, with maximum vertical and horizontal displacements of about 3 feet. The cracking and movements observed in this area were coincident 

with topographic features associated with landsliding. Observations in exploration trenches indicated that these cracks 

split bedrock materials and that the fissures existed prior to 

the earthquake. 

On the left abutment, nearly continuous cracking was observed about 400 feet along an access road from the dam 

crest to the upstream inlet structure. The cracking generally 

paralleled the slope, with vertical scarps up to 14 inches in 

height, and was subsequently determined by exploratory 

trenching to be the expression of a shallow landslide in loose 

material overlying graywacke and shale bedrock that had 

experienced previous, lesser movements. This landslide did 

not threaten the inclined inlet. 

In addition to several small slumps and landslides around 

the rim of the reservoir, a set of aligned fissures was observed 

after the earthquake on the right side of the reservoir. These 

fissures developed along a topographic bench for a distance 

of about 1,500 feet upstream of the dam (fig. 3). In general, 

the fissures appeared to correspond to slumping of the ground 

toward the reservoir, with vertical scarps up to 3 feet being 

measured. The dam owner's geotechnical consultant 

excavated borrow and exploration trenches in this area and 

concluded that the aligned cracking resulted from shakingTable 6.-Characterization of Austrian Dam/ill materials 

Engineering property Range Mean 

USCS Classification SC, GC, CL 

Gradation: > No. 4 (%) 26.0 - 71.5 46.3 

Gradation: < No. 200 (%) 16.0 - 43.7 3 1.8 

Specific Gravity, Gs 2.60 - 2.78 2.70 

Liquid Limit 28 - 32 3 1 

Plasticity Index 11 - 15 13 

Wr (as compacted -1950, %) 9.5 - 19.5 14.5 

-yd (as compacted -1950, pcf) 107.5 - 132.0 121.1 

yd (in situ-1989, pcf) 121.3-131.6 126.6 

C- (psf-1 0 

(ff (degrees) 44 

(PS~) 290 

<p (degrees) 21 

K2ma 106 - 128 122 
PERFORMANCE OF THE BUILT ENVIRONMENT 

induced settlement of loose, clayey fill overlying a steep 

bedrock surface and that the linear orientation of the fissures 

was most likely a result of excavation and/or shaping of the 

ground during the initial dam construction (Wahler 

Associates, 1990). 

There remains, however, some disagreement as to the cause 

of the aligned fissures. In a study by Aydin and others (1992), 

another set of cracks, located approximately 500 feet upslope 

of the aligned fissures, was mapped. These higher cracks 

had offsets commonly between 2 and 8 inches, and the study 

concluded that they were associated with sympathetic 

tectonic movements on the primary strand of the Sargent 

fault during the earthquake. That study went on to suggest 

that the aligned fissures on the bench along the reservoir 

might be associated with movements along another strand 

of the Sargent fault. 

Figure 5 presents horizontal and vertical movements for 

crest monuments measured just prior and after the 

earthquake. These measurements indicated that the dam crest 

settled over 2.5 feet along most of its length. In addition, the 

right end of the dam appeared to move downstream 

horizontally 1.5 feet relative to the left end. Since the survey 

monuments were not tied into a stationary benchmark, 

absolute displacement vectors cannot be calculated. 

Movements and damage near the toe of the dam suggest 

that the dam moved primarily downstream. 

The strong shaking and ground movements produced 

extensive cracking in the crest of the dam near both 

abutments. Near the left end of the crest, predominantly 

transverse cracks in the embankment were up to about 8 

inches in width and 10 feet in depth. However, the most 

pronounced cracking near the left end of the dam occurred 

along the embankment-foundation contact, where vertical 

scarps up to 16 inches were observed at the surface. 

Subsequent explorations traced open cracks varying from I/ 

8 inch to as much as 1 112 feet in width along this contact to 

depths of up to about 27 feet. The explorations suggested 

that the cracks were within old landslide deposits that had 

been left in place on the upper part of the abutment. The 

possibility that some undisclosed separations might have 

occurred at even greater depths led to a subsequent remedial 

grouting program to further examine the contact at greater 

depths. 

Particularly severe cracking occurred where the right end 

of the dam abuts onto the concrete spillway. This area of the 

spillway includes an entrance section from the unlined 

approach channel, the weir, and a transition section which 

converges to the chute section. The entrance section includes 

a wing wall extending upstream along the left side of the 

approach channel, and a "return" wall coming back in a 

downstream direction, presumably intended to prevent scour 

around the upstream end of the wing wall. The two walls 

thus form a "U" pointing downstream, with the right end of 

the dam abuting on the outside of the return wall and also 

backfilling the space between the two walls. Up to 9-inchwide diagonal cracks occurred in the embankment between 

the walls. Separations up to 10 inches also developed between 

the inner faces of the walls and the enclosed embankment. 

A separation along the outside face of the return wall 

extended to a depth of about 23 feet (nearly the base of the 

wall). This would have been particularly dangerous had the 

reservoir been full, as the normal amount of freeboard was 

only about 15 feet. 

Immediately downstream of the spillway entrance, where 

the dam abuts the left wall of the transition section, several 

cutoff collars on the outer (embankment) side of the wall 

were sheared-off due to upstream-downstream movement 

of the adjoining embankment relative to the wall. 

Extensive damage also occurred in the transition and chute 

sections of the spillway walls. Up to 314-inch-wide tension 

cracks were observed in the walls and slab of both sections. 

The cracking in the slab of the chute section was somewhat 

regularly spaced at about 2 to 4 feet, normal to the axis of 

the spillway. The 80 feet of chute and transition section 

farthest upstream was found to have elongated a total of about 

7 inches. In the transition section, tension cracks generally 

paralleled the trend of cracking which occurred in the natural 

slope above the spillway. Up to 1-inch-wide separations were 

observed between the bottom of the slab and its highly 

weathered bedrock foundation along the left wall of the 

transition section. Subsequent grouting through the slab 

indicated that these voids developed existed primarily 

beneath the walls, but not beneath the central part of the 

slab. This suggested that the spillway walls had been 

subjected to some sort of rocking action. At the downstream 

end of the chute, the concrete wingwall on the left side was 

rotated and torn away from the left wall, while the right wing 

wall could not be found. 

In addition to the damage to the crest and spillway, a series 

of roughly parallel longitudinal cracks formed in the 

upstream and downstream faces of the dam. The initial 

widths of the cracks were relatively small, but ultimately 

widened within a few weeks to as much as a foot in places. 

These time-dependent movements may have resulted from 

major aftershocks, rainfall during this period, soil creep, 

and/or pore-pressure dissipation and consolidation. Seven 

days after the main shock, the four longitudinal cracks which 

formed in the upper 50 feet of the upstream face were 

approximately 5 to 15 feet deep and 1 to 4 inches wide. A 

number of longitudinal cracks which were 3 to 8 feet deep 

and 2 to 6 inches wide formed in the downstream face. The 

majority of the longitudinal cracks in the downstream face 

were located near the crest, although some limited cracking 

also occurred near the toe of the dam. In addition, there 

was some minor bulging of the downstream toe. Exploration 

trenches indicated that the longitudinal cracks generally 

dipped steeply toward the dam crest on both embankment 

slopes. Hence, the cracks did not appear to result from slope 

instability, but rather from settlement and rearrangement 

of the earth embankment. 
PERFORMANCE OF EARTH DAMS DURING THE LOMA PRIETA EARTHQUAKE 

- Original alignment 

*-*Â¥**- Oct 16, 1989 alignment 

**X-uu Qct 19, 1989 alignment 

0.0 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0 

L STATION (IN 100 FT) R 

Original alignment 

<^fr*-* Oct 16, 1989 alignment 

Oct 19, 1989 alignment 

0.0 1 .O 2.0 3.0 4.0 5,O 6.0 7.0 8.0 

L STATION (IN 100 FT) R 

Figure 5.-Horizontal (A) and vertical (B) movements of crest monuments at Austrian Dam (from Seed and others, 1990). 
Dl2 PERFORMANCE OF THE BUILT ENVIRONMENT 

The movements of the embankment were concluded to be 

due to general settlement and spreading of the fill during the 

strong earthquake shaking, followed by subsequent 

downslope creep. Figure 6 presents piezometer measurements 

made for 5 standpipe piezometers installed prior to the earthquake. These instruments recorded pore-pressure increases 

between 12 and 54 feet of water. Possibly related to the porepressure increases is the fact that some of the trenches excavated across the longitudinal cracks near the crest encountered free water at elevations considerably higher than reservoir levels at the time of the investigation (Wahler Associates, 1990). 

The largest pore-pressure increase, 54 feet of water, occurred in piezometer P-1 , with its tip located within the downstream "pervious" zone near the bedrock contact. The 

standpipe for this instrument was found to be significantly 

deformed between elevations 955 and 960 (about 25 to 30 

feet above the bedrock contact). The standpipe for piezometer P-6, which measured an increase in hydraulic head of 

about 49 feet of water, was also found to be deformed between elevations 1,017 and 1,041 feet (about mid-height of 

the embankment). These deformations were suggestive of 

earthquake-induced internal movements corresponding to 

lateral spreading of the embankment. 

The repair of the earthquake damage consisted of (1) excavating and recompacting the fill in the areas of extensive 

cracking, (2) placing a zoned fill with chimney and blanket 

drains in the crest fill at both embankment ends, (3) excavating and recompacting the upstream face of the fill to create 

an impervious blanket, (4) epoxy grouting the cracking in 

the spillway and cement grouting voids beneath the spillway slabs, and (5) grouting of the rock at the left abutment 

contact with the fill. 

This repair successfully remediated the earthquake-induced 

damage to the embankment and was accomplished within 8 

weeks following the earthquake. Details of the repair can be 

found in Wahler Associates (1990). Due to the severity of damage to the concrete spillway and concern about potential future landslides in the right abutment, a new spillway was later 

constructed on the left abutment of the dam in 1993-1994. 

LOMA PRIETA EARTHQUAKE 

- Surface water due to - 

construction activities 

-.-a -.-.-. 

- 

-----".- 

I 

I 

i 

< 

- 1 

I - 

, :.*. .................................................... ..................................... P-2 

I : 

................................... P-2 / I:/\ -^ -- 

------- -9- L --L 

---------- - ----I -t^----&..---a A --. ^k*----- P- 4 - 

SEPTEMBER 1 OCTOBER 1 NOVEMBER 1 DECEMBER 1 JANUARY 1 FEBRUARY A 

1989 1990 

Figure 6.-Piezometer measurements for preearthquake piezometers at Austrian Dam (from Wahler Associates, 1990). 
1 PERFORMANCE OF EARTH DAMS DURING THE LOMA PRIETA EARTHQUAKE D 13 

LEXINGTON DAM 

Lexington Dam is a 205-foot-high dam located about 6 

miles downstream of Austrian Dam and about 2 miles from 

the fault rupture associated with the earthquake (fig. 2). The 

dam was completed in 1953 as a zoned earth structure having a relatively thick sandy and gravelly clay core that is 

supported by upstream and downstream random shell zones 

of clayey sands and gravels. The dam also has relatively flat 

upstream (5.5: 1) and downstream (3: 1) slopes. A plan view 

and cross section are shown in figure 7. The embankment 

material properties are summarized in table 7. At the time of 

the earthquake, the reservoir was about 100 feet below the 

crest of the dam. Previous summaries of damage were presented by Bureau and others (1989), Seed and others (1990), 

and in the studies by R. L. Volpe & Associates (1990a). 

Lexington Dam was instrumented with strong-motion instruments on the left abutment, left crest, and right crest. 

These accelerographs recorded transverse peak accelerations 

of 0.45,0.39, and 0.45 g, respectively. This shaking was composed of about 6 to 7 s of relatively strong long- period motion. The left abutment or "bedrock" peak acceleration is 

within the range predicted by appropriate strong-motion attenuation relationships for a site approximately 2 miles from 

the nearest point on the fault rupture surface for a Ms = 7.1 

event, but is a bit lower than the mean or expected value 

based on such relationships. In addition, there appears to 

be some spectral acceleration amplification at lower frequencies (0.9 to 1.2 Hz). This low frequency amplification may 

indicate that the recorded "bedrock" motion may have been 

affected by local topographic or geologic conditions. 

The strong ground shaking produced transverse cracking 

on both the upstream and downstream sides of both abutments, oblique cracking on the crest about 150 feet in from 

the left abutment, longitudinal cracking on both the upstream 

and downstream slopes of the dam, and cracking of an access road on the right abutment upstream of the dam. The 

cracks, which were fairly isolated, were commonly less than 

314 of an inch wide, and trenching indicated that they only 

extended to depths generally between 2 and 7 feet (R. L. 

Volpe & Associates, 1990a). The maximum earthquake-induced crest deformations were approximately 0.85 feet of 

vertical settlement, and 0.25 feet of lateral displacement in 

the downstream direction (R. L. Volpe &Associates, 1990a). 

An old slope indicator casing was found to have raised from 

beneath the crest to over 3 inches above the crest due to the 

embankment settling around it. The earthquake shaking and 

ground movements produced extensive cracking in the bridge 

abutment at the left abutment and ruptured a buried water 

line near the crest of the dam. 

About 6 weeks after the earthquake, a relatively large seepage area developed high up on the downstream face of the 

dam. The seepage area was about 170 feet long and 35 feet 

wide and oriented at an oblique angle with the axis of the 

dam. This seepage area was really more of a wet or damp 

area and never really flowed water. Although the cause of 

the seepage area is not definitively known, one explanation 

that has been offered is the fact that old exploration holes 

extending into the rock foundation lie within the area and 

that these old borings could have been acting as relief wells 

for earthquake-induced pore pressures within the lower portions of the embankment and bedrock (R. L. Volpe & Associates, 1990a). Another possible explanation is that the fill is relatively impervious at this elevation and that any surface water 

that infiltrates the dam becomes perched at this level. 

The repairs made to the dam consisted of trenching the 

cracked areas to depths ranging between 3 and 7 feet and 

compacting the excavated soil back into the trenches (R. L. 

Volpe & Associates, 1990b). 

GUADALUPE DAM 

Guadalupe Dam is a 142-foot-high dam located about 6 

miles from the Loma Prieta fault-rupture zone, and it probably experienced peak ground accelerations between 0.4 and 

0.45 g (fig. 2). The dam was completed in 1935 as a rolled 

earth structure with an upstream facing of concrete panels 

for erosion protection. In a manner similar to that described 

for Austrian Dam, the embankment is apparently nearly homogeneous, as the selective borrowing to create upstream 

"impervious" and downstream "pervious" zones did not appear to be completely successful in creating distinctly different zones. In 1972, an upstream buttress was added to the 

dam to improve drawdown stability. A plan view and cross 

section are shown in figure 8. At the time of the earthquake, 

the reservoir was about 78 feet below the crest of the dam; 

however, the reservoir had been full up to about 3 months 

before the earthquake, and it is assumed that the upstream 

shell materials were nearly saturated at the time of the earthquake. Previous summaries of damage were presented by 

Bureau and others (1989), Seed and others (1990), and in 

the studies by R. L. Volpe & Associates (1990a). 

The earthquake induced up to 0.64 feet of settlement and 

0.15 feet of lateral displacement in the upstream direction as 

measured on the crest. Minor transverse cracking developed at 

the crest at both abutment contacts along with minor longitudinal cracking on the crest. The principal damage was to the upstream slope, where the upper portion of the buttress fill developed longitudinal cracking. Shortly after the earthquake, these 

cracks were observed to have a maximum width of less than 1 

inch and extended across the entire face of the dam. About 5 

weeks later, the cracks had widened to about 4 inches and extended 

to a depth of about 5 feet (R. L. Volpe &Associates, 1990a). 

These cracks may have been caused by concentrations of 

dynamic stresses induced by the change in slope geometry. 

Alternatively, they may have resulted from possible past 

settlements caused by the placement of the berm. These past 

settlements may have created preexisting cracks which surfaced only after the development of strong ground motion. 
UPSTREAM TOE 

PERFORMANCE OF THE BUILT ENVIRONMENT 

1 SPILLWAY 1 

DOWNSTREAM TOE 

ORIGINAL GROUND 

Figure 7.-Plan view (A) and cross section (5) of Lexington Dam. 
PERFORMANCE OF EARTH DAMS DURING THE LOMA PRIETA EARTHQUAKE Dl5 

Table 7.-Characterization of Lexington Dam fill materials 

Engineering CORE SHELL 

Property (Zone 2) (Zones 1 & 4) 

Depth (ft) 0-80 >80 

Classification SC, CL-CH 

Gradation: > No. 4 (%) 13-30 0-2 

Gradation: < No. 200 (%) 29-52 86-97 

Specific Gravity, Gs 2.67 - 2.73 

Liquid Limit 31 - 37 

Plasticity Index 14 - 18 

Wc (in situ range 11.2-17.7 21.5-30.6 

and avg., %) (14.4) (25.6) 

y (in situ range 117.5-131.5 92.5-102.2 

and avg., pcf) (120.9) (96.6) 

ydmax, wcopt 

(20,000 ft-lblft3) 

c' sf) 400 0 

(ff (degrees) 36 25 

C (psf) 1200 0 

<b (degrees) 18 17 

K2ma.x 100-150 -50 

Vsmm (ftlsec) 1200-1 600 1400-1 600 

SC GC, CL 

0-50 

'1 8-97 

2.73 

33 - 39 

14 - 24 

9.4-26.5 

(15.3) 

95.2-134.8 

(1 19.8) 

131.7 pcf 

8% 

0 

36 

0 

22 

-100 

1400-2200 

Notes: Only one sample had less than 13 percent gravel. 

For 70 percent of the samples, <35 percent finer than No. 200. 

Average dry unit weight was about 6 pcf lower at depths 

of 0-40', and 4 pcf higher at depths >40'. 

The earthquake also caused the concrete panels on the 

upstream face above the berm to pound against each other, 

resulting in cracking and spalling in about 10 percent of the 

panels (R. L. Volpe & Associates, 1990a). 

The repairs made to the dam consisted principally of excavating a 70-foot band of material (as measured along the slope, 

parallel to the crest) at the top of the upstream buttress to a 

depth of about 6 feet. The excavated material was temporarily 

stored in order to allow the material to dry to an acceptable 

water content and then recompacted into place. The cracks in 

the crest at the abutment contact were excavated to about 3 to 

4 feet and the excavated material was recompacted back into 

place (R. L. Volpe & Associates, 1990b). 

NEWELL DAM 

Newel1 Dam is a 182-foot-high dam located about 6 miles 

from the Loma Prieta fault-rupture zone, and it probably experienced peak ground accelerations between 0.4 and 0.45 g (fig. 

2). The dam was completed in 1960 as a zoned earth and rockfill 

dam generally composed of clayey zones except for an upstream zone of dirty rockfill. At the time of the earthquake, the 

reservoir was about 49 feet below the crest of the dam. Previous summaries of damage were presented by Bureau and others (1989), Seed and others (1990), and Creegan (1990). 

Although the earthquake did not induce significant crest 

movements, a longitudinal crack was found on the 3: 1 upstream 

slope running the entire width of the dam face at about the 

spillway elevation. This crack was generally between 1 and 9 

inches in width. Trenching explorations of the crack indicated 

that it was formed by tension within the zone 2 rockfill. The 

trenches, however, only extended to a maximum depth of 10 

feet, and the crack, which was about 1 inch in width at the 

bottom of the trenches, extended farther to greater depths. There 

were also other minor cracks at various locations. Seepage through 

the dam and abutments, measured at the downstream toe, was 

also found to have increased from a normal 8 gpm to 41 gpm, 

but remained clear (Creegan, 1990). By early December 1989, 

the seepage had decreased back down to about 17 gpm. 

The development of the longitudinal cracking was theorized to have resulted from settlement of the dirty rockfill 

upstream shell relative to the rest of the dam. This material 

consisted of quarried sandstone and shale and was placed in 

5-foot lifts and compacted by sluicing. The other clayey zones 

were placed in thin lifts and compacted to about the maximum Standard Proctor dry density. Consequently, the upstream rockfill zone was relatively loose in comparison with 

the other zones in the embankment. There were indications 

that some of the cracking ran along the interface of the zone 

2 rockfill and the clayey core. Further evidence of settlement of the zone 2 material was found at the bell toggle joints 

along the sloping intake tower where the embankment 

seemed to have pulled away from the structure by about 1 to 

3 inches in a downstream direction (Creegan, 1990). 

The repair of the longitudinal cracks consisted of using a 

large backhoe to excavate to a depth of about 6 feet along 

the alignment of the crack. The trench was then backfilled in 

18-inch lifts with each lift being compacted with a vibrating 

shoe on the backhoe. The upstream slope was then rolled 

with a vibratory roller (Creegan, 1990). 

ELMER J. CHESBRO DAM 

The Elmer J. Chesbro Dam is a 95-foot-high embankment 

located about 8 miles from the Loma Prieta fault-rupture 

zone, and it probably experienced peak ground accelerations 

between 0.4 and 0.45 g (fig. 2). As for many of the embankment dams in this area, the embankment is a nearly homogeneous compacted fill, as selective borrowing to create upstream "impervious" and downstream "pervious" zones do 

not appear to be entirely successful in creating distinctly different materials. The upstream slope varies between 2: 1 and 

3: 1. The downstream slope is about 2: 1, but is fitted with a 

45-foot-wide berm at about half the height of the dam. At 

the time of the earthquake, the reservoir was about 69 feet below the crest of the dam. Previous summaries of damage were 

presented by Bureau and others (1989), Seed and others (1990), 

and in the studies by R. L. Volpe & Associates (1990a). 

Surveys of crest monuments showed that the earthquake 

induced up to 0.37 feet of settlement and 0.05 feet of lateral 

displacement in the upstream direction. The main area of 

cracking that developed at this dam occurred as longitudinal 

cracking near the upstream edge of the crest. This cracking 

extended about 240 feet and had a 4-inch width together with 
Dl6 PERFORMANCE OF THE BUILT ENVIRONMENT 

a 4-inch vertical offset, with the upstream side of the crack 

being on the down side. There was also minor transverse 

cracking on the right abutment at both the crest area and 

near the spillway entrance (R. L. Volpe &Associates, 1990a). 

Trenching explorations indicated that the longitudinal cracking on the upstream edge of the dam extended to a maximum 

depth of about 8 feet and was related to settlement of an upstream slope-protection zone. The transverse cracking on the 

crest at the right abutment was estimated to have a maximum 

depth of 4.5 feet (R. L. Volpe & Associates, 1990a). 

The repairs to Chesbro Dam consisted of excavating the 

upper 10 to 12 feet of the upstream slope and replacing this 

material with compacted, imported materials. The cracking 

on the right abutment was repaired by trenching to a depth 

of about 5 feet and recompacting the excavated material back 

into place (R. L. Volpe & Associates, 1990b). 

I 

UPSTREAM TOE 

ILLWAY 

30' 

CONCRETE FACE 7 

DOWNSTREAM TOE 

SCALE (feet) 

o-oo 

Figure 8.-Plan view (A) and cross section (B) of Guadalupe Dam. 
PERFORMANCE OF EARTH DAMS DURING THE LOMA PRIETA EARTHQUAKE Dl7 

VASONA PERCOLATION DAM 

The Vasona Percolation Dam is a 34-foot-high embankment located about 5.5 miles north of the Loma Prieta faultrupture zone, and it probably experienced peak ground accelerations between 0.35 and 0.4 g (fig. 2). The dam is composed of two embankments flanking a central concrete spillway. The two embankments were completed in 1935 and contain an upstream zone of rolled "fine" material together with 

downstream zones of coarse and random materials. Beneath 

the upstream "fine" material there is a puddled clay cutoff trench 

which extends 25 feet to an impervious foundation material. 

At the time of the earthquake, the reservoir was about 12 feet 

below the dam crest (R. L. Volpe & Associates, 1990a). 

The dam developed maximum settlements of about 0.16 

feet and about 0.09 feet of lateral movement in the downstream direction (R. L. Volpe &Associates, 1990a). The major 

damage consisted of longitudinal cracking along the crest of 

the right embankment. This cracking was intermittent, but 

extended along the entire length of the embankment. These 

cracks were up to 314 of an inch in width, and trenching 

indicated a maximum depth of about 5.5 feet. There was also 

limited longitudinal cracking on the crest of the left crest 

embankment, generally between 118 to 112 inches in width. 

Minor transverse cracking, generally less than 118 of an inch 

in width, was found along the spillway contact from the crest 

centerline to the downstream toe. Cracking was also noted 

along the spillway wingwall at the downstream toe of the 

dam and along the parapet wall along the crest of the dam. 

Longitudinal cracking of the crest had previously been 

noted following the August 8,1989 earthquake (M=5. I), now 

considered a possible foreshock of the Loma Prieta earthquake. In some locations, the longitudinal cracks observed 

EARTHQUAKE-INDUCED CRACKS 

IN DAM FILL 

following the Loma Prieta earthquake were in the same locations as those from the earlier earthquake. 

The repair principally consisted of removing the upper 4.2 

feet of the right embankment and recompacting the material 

back into place (R. L. Volpe & Associates, 1990b). 

SODA LAKE DAM 

Soda Lake Dam consists of four embankments that are used 

to retain granite tailings from a nearby quarrying operation. 

The dam had been enlarged in 1978 to give a maximum height 

of 35 feet for the main or south embankment. The dam is located approximately 5.5 miles from the Loma Prieta fault-rupture zone, and it probably experienced peak ground accelerations between 0.3 and 0.35 g (fig. 2). Although the reservoir 

was nominally empty at the time of the earthquake, most of the 

reservoir volume had been filled with granite tailings. 

Although the reservoir may have been considered empty, 

portions of the granite tailings were evidently saturated, as 

they appeared to have liquefied in several places. Evidence 

of liquefaction was in the form of numerous sand boils developed within the reservoir deposits and the fact that the 

reservoir sediments settled about 2 to 3 feet relative to the 

embankments and reservoir rim. One reconnaissance team 

found the sand boils still ejecting water and sand 72 hours 

following the main shock (J. Egan, oral commun., 1993). 

No damage was found at the north, south, or east embankments. However, the 10-foot-high West Saddle Dam was 

found to have an extensive pattern of cracking at and below 

the crest. As shown in figure 9, the cracking was in the pattern of an arc and seemed to identify a wedge of material 

that had developed into a minor slump towards the reserLIMIT OF TAILINGS PRIOR 

TO DAM CONSTRUCTION \ 

CREST OF DAM 

50 100 feet SCALE 7-1 

Figure 9.-Plan view of the Soda Lake west saddle dam showing earthquake-induced cracking. 
D 18 PERFORMANCE OF THE BUILT ENVIRONMENT 

voir. These cracks had typical widths of about 1 to 4 inches 

and could be probed to depths of at least 6 feet. The cracks 

also had vertical offsets of about 1 inch with the reservoir 

side of the crack being lower than the downstream side. 

The cause of the cracking and/or slumping of the West 

Saddle Dam was theorized by engineers within the California Division of Safety of Dams to be a result of having left 

potentially liquefiable tailings deposits in the foundation 

beneath this embankment during its construction in 1978. 

As shown in figure 10, prior to construction of the West 

Saddle Dam, preexisting tailings extended from the reservoir to about the centerline of the dam to be built. These 

tailings were supposed to have been removed. However, if 

they were not completely removed, then liquefaction of tailings remnants within the foundation might explain the cracking and movements observed following the earthquake. Supporting this theory is the fact that the shape and location of 

the cracked wedge appears to match the contour of the limits 

of preexisting tailings in the foundation (figure 9). 

As a result of the cracking/slumping in the West Saddle 

Dam, the dam was not considered safe to store water, and 

the owner was required to take the dam and reservoir out of 

service. 

MILL CREEK DAM 

Mill Creek Dam is a 76-foot-high dam with a complicated history. It was originally built in 1889 as a timber crib 

dam, with a substantial hydraulic fill zone placed upstream 

of the timber crib portion. There is apparently no information available concerning its status and performance following the 1906 earthquake. In 1932, additional fill was placed 

upstream to give the upstream face a 3:l slope. In 1947, 

part of the timber crib dam burned. Part of the burned timber crib zone was removed, leaving a downstream slope of 

about 1.5: 1. In 1957, a large sinkhole formed in the upstream 

face above corroded sections of the original 14-inch outlet 

pipe. The repair for the sinkhole consisted of excavating a 

large triangular trench to remove portions of the corroded 

pipe and then recompacting the excavated materials back 

into the trench. A new outlet consisting of a 12-inch pipe 

was also installed at this time. Figures 11 and 12 present 

sections illustrating the different materials present in the 

dam. 

Mill Creek Dam is located approximately 12 miles from 

the Loma Prieta fault-rupture zone, and about 17 miles from 

the epicenter. The dam may have experienced peak ground 

accelerations of up to about 0.3 g (fig. 2). At the time of the 

earthquake, the reservoir was about 16 feet below the crest. 

Despite the questionable materials left within it, the dam 

performed well during the earthquake. The only damage 

observed was minor cracking at the crest, which became obscured during the rains that followed a few days after the 

earthquake. 

The good performance is particularly surprising due to the 

presence of the sandy and silty hydraulic fill left in place 

beneath the 1957 repair trench. Piezometers indicated that 

these soils were saturated prior to the earthquake. Penetration tests made during investigations conducted in 1987 and 

1992 indicated that these soils are very loose [Standard Penetration Test (Nl)eo,cs of the hydraulic fill is equal to about 10 

blows per foot]. 

A Standard Penetration Test-based analysis of the dam was 

performed by Wahler Associates (1992) for the Loma Prieta 

earthquake using an estimated peak acceleration in bedrock 

of 0.17 g. The results indicated that liquefaction would be 

triggered in the hydraulic fill, resulting in a flow slide of the 

upstream slope during the Loma Prieta earthquake. A backanalysis was also performed to determine the value of residual strength of the hydraulic fill which would have been 

needed to prevent the calculated onset of a flow slide and 

result in computed deformations in agreement with the actual observed seismic displacements. It was determined that 

a residual strength of about 600 psf would have been needed. 

However, based on the Standard Penetration Test results, the 

estimated actual in-situ residual strength would only have 

been about 300 psf. 

A possible explanation for the discrepancies between observed and back-calculated performances is that the site acceleration could have been significantly less than the estimated 0.17 g peak value. Low cyclic loadings might not have 

caused liquefaction of the hydraulic fill and would explain 

the lack of damage to the dam. There were no seismic instruments at the site, and there were no physical indications 

that liquefaction had occurred. 

Although the dam performed well during the 1989 Loma 

Prieta earthquake, design modifications to the existing spillway were developed to restrict the reservoir surface to an 

operating level which would reduce the risk of sudden release of the reservoir during a future large earthquake. 

LEROY ANDERSON DAM 

Leroy Anderson Dam is a 235-foot dam located about 13 

miles from the Loma Prieta fault-rupture zone. The dam was 

completed in 1950 and is composed of central core zones of 

compacted sandy clay and clayey sand that are flanked by 

shell zones of rockfill. The rockfill was placed by end-dumping from trucks into 10 to 25-foot-high lifts that were compacted by sluicing with water (fig. 13). 

The dam is well instrumented with seismographs, which 

indicated a peak horizontal ground acceleration of about 0.26 

g at the base together with a crest horizontal peak acceleration of about 0.43 g. There were several other instruments 

placed on the crest and downstream face which recorded peak 

horizontal accelerations of between 0.1 and 0.4 g. 

This dam also experienced strong shaking during the 

nearby 1984 Morgan Hill earthquake (Mr=6.2), with a peak 
1 PERFORMANCE OF EARTH DAMS DURING THE LOMA PRIETA EARTHQUAKE 

20 feet 170 i 1 170 

1 1 ELEVATION 165 feet 

Figure 10.-Cross section of the Soda Lake west saddle dam. 

COMPACTED TRENCH 

RESERVOIR ELEVATION 

ELEVATION 1487 feet 

SACK 

AP 

Figure 11 .-Cross section of Mill Creek Dam . 
PERFORMANCE OF THE BUILT ENVIRONMENT 

1' 0 50 loo DOWNSTREAM TOE 

fi CREST EL. 640' 

Figure 13.-Plan view (A) and cross section (B) of Leroy Anderson Dam. 

acceleration at the downstream toe of about 0.41 g and a 

corresponding horizontal peak acceleration of 0.63 g at the 

crest. Although of larger peak amplitude, the duration of 

strong shaking was less during the 1984 earthquake than 

during the 1989 event. 

Both the 1984 and 1989 earthquakes produced similar 

patterns of distress in the embankment. In 1984, two parallel, and relatively shallow, longitudinal cracks formed in the 

compacted fill along the crest approximately overlying the 

buried contacts between the central clayey core and the upstream and downstream rockfill shell zones. These cracks 

were generally less than an inch in width and are thought to 

have been the result of settlement of the shell zones relative 

to the core zones. Between 1984 and 1989, the embankment 

crest was raised about 5 feet and a small sliver fill was placed 

in a localized area on the downstream edge of the crest. The 
PERFORMANCE OF EARTH DAMS DURING THE LOMA PRIETA EARTHQUAKE D2 1 

5-foot crest raise was placed to provide additional head for a 

modified spillway, and the sliver fill was placed to provide 

space for an instrumentation vault. 

The main damage following Loma Prieta was a longitudinal crack on the downstream half of the crest approximately 

above the contact between the shell and core zones. In addition, the longitudinal cracking extended to and around the 

base of the vault on top of the sliver fill. These cracks were 

generally less than 314 of an inch in width. A second set of 

longitudinal cracks was also observed at both edges of the 

paved crest road along the bases of guard rail posts. This 

latter set of cracks were as much as 1.5 inches in width. 

The maximum settlements of crest monuments following 

Loma Prieta were less than 0.04 feet on the upstream edge 

and about 0.13 feet on the downstream edge. Only minor 

repairs were considered necessary for the crest cracking (R. 

L. Volpe & Associates, 1990a). 

SAN JUST0 DAM 

San Justo Dam is a 135-foot dam located about 17 miles 

from the Loma Prieta fault rupture zone (fig. 2). The dam is 

relatively new and was completed in 1986 as a zoned earth 

and rockfill embankment (figs. 14 and 15). Although no significant damage was observed following the earthquake, this 

dam provides an outstanding opportunity to investigate the 

dynamic response characteristics of earth dams. This is because the dam is heavily instrumented with strong-motion 

sensors. Furthermore, one of the sensors was embedded in a 

borehole approximately 62 feet below the crest of the dam, 

providing one of the few opportunities to calibrate response 

analyses with the internal motions of an embankment dam. 

Figures 14 and 15 illustrate the peak accelerations recorded 

at different locations at San Justo Dam. The transverse horizontal peak acceleration recorded at the downstream toe was 

about 0.26 g; whereas the transverse horizontal peak acceleration recorded at the crest was about 0.40 g, an amplification of about a factor of 2. However, the interior sensor at 

about the mid-height of the dam indicated a transverse peak 

acceleration of only about 0.26 g. This would suggest that 

most of the transverse amplification took place within the 

upper half of the dam. 

The strong-motion sensors also suggest that longitudinal 

peak horizontal accelerations were amplified by about a factor of 3, but that the longitudinal amplification took place in 

the lower half of the embankment. 

HAWKINS DAM 

Hawkins Dam is a 72-foot-high hydraulic fill that was built 

slowly between 19 12 and 193 1. It is located about 2 1 miles 

from the Loma Prieta fault-rupture zone, and it probably 

experienced peak horizontal ground accelerations of about 

0.2 to 0.25 g (fig. 2). Although the reservoir was empty at 

the time of the earthquake, the dam experienced minor longitudinal cracking on the crest and on the upstream slope. 

The dam was investigated by Lee and Roth (1977) during 

the performance of a seismic stability evaluation. This study 

found that the dam was generally composed of a fairly homogeneous and impervious gravel-sand-clay fill, together 

with a small zone of homogeneous clay in the central portion of the embankment (fig. 16). Because it was built slowly, 

with only about 2 to 3 feet of material added each year, there 

was time for consolidation and desiccation of the fill to take 

place. The study by Lee and Roth (1977) concluded that the 

dam would have adequate stability for even a magnitude 8+ 

earthquake on the San Andreas fault. Unfortunately, because 

the reservoir was empty at the time of the earthquake, an 

opportunity to verify this evaluation was lost.] 

PERFORMANCE OF EARTH DAMS AT 

GREATER DISTANCES 

In addition to the dams located in relatively close proximity to the earthquake, the performance of a few dams located 

at greater distances is also of interest. 

CALAVERAS DAM 

Calaveras Dam is a 210-foot-high hydraulic fill dam that 

was completed in 1925. During construction, it developed a 

large slide due to static liquefaction. Although the larger 

portion of the embankment is composed of hydraulic fill, 

the upper portion was built with a rolled clayey core flanked 

by rockfill shell zones. In 1974, a rockfill buttress was added 

to the upstream slope to increase freeboard and to provide 

increased stability. 

The dam is located approximately 24 miles from the Loma 

Prieta fault-rupture zone, and it probably experienced peak 

ground accelerations between 0.1 and 0.15 g (fig. 1). At the 

time of the earthquake, the reservoir was about 70 feet below the crest of the dam. 

The dam experienced no significant damage. Reported 

incidents consisted of minor changes in piezometers, a temporary 25 to 30 percent increase in seepage, and the muddying up of seepage water attributed to a stirring up of silt by 

the shaking of the earthquake. 

SAN LUIS DAM 

San Luis Dam is a 313-foot-high dam that was completed 

in 1967. It is composed of a homogeneous section of compacted clayey material with internal drains and is flanked 

with slope protection zones. In 1981, a portion of the upstream slope developed a large slide following a severe draw- 
^' / SCALE 1 

0 200 400 feet I I 

Figure 14.-Plan view of San Justo Dam. 

CREST ELEVATION 508.5 feet 

SCALE ? 1 00 200 feet 

Figure 15.-Maximum (A) and Left Quarter (5) sections of San Justo Dam. 
PERFORMANCE OF EARTH DAMS DURING THE LOMA PRIETA EARTHQUAKE D23 

0 50 feet 

SCALE 1-J 

Figure 16.Ã‘Cros section of Hawkins Dam (from Lee and Roth, 1977). 

down of the reservoir. Although the slide began below the 

crest, the slide propagated far enough back to take out a portion of the crest. The slide was found to involve slopewash 

material left in place in the foundation. The dam was repaired by leaving the slide material largely in place, buttressing the slide with a large upstream berm, and rebuilding 

the crest section. 

San Luis Dam is located about 34 miles from the Loma 

Prieta fault-rupture zone (fig. 1). At the time of the earthquake, the reservoir was less than half full. Strong motion 

sensors at two locations along the downstream toe recorded 

peak horizontal accelerations of 0.06 and 0.10 g. The corresponding locations on the crest recorded peak horizontal accelerations of 0.25 and 0.17 g. Despite experiencing moderate levels of ground motion, the dam experienced no significant damage. Two days after the earthquake, small longitudinal hairline cracks were found on the crest in the vicinity 

of the slide. These cracks appeared to be largely preexisting 

cracks which may have widened slightly following the earthquake. Longitudinal hairline cracks were also found on the 

crest at other locations. The largest crack was longitudinal 

and had a width of less than 114 of an inch. It was located 

near the right abutment, where the height of fill was only 

about 5 feet. No repairs were needed at this dam as a result 

of the earthquake. 

O'NEILL DAM 

O'Neill Dam stores a reservoir that serves as the forebay 

to San Luis Dam. It was also completed in 1967 and is a 

homogeneous earth dam with an internal downstream blanket drain and exterior slope protection zones. It has a maximum height of about 70 feet (fig. 17). 

Recent studies have indicated that the alluvial foundation 

contains in some locations a layer of potentially liquefiable 

sand lying immediately beneath a surface clay cap. Corrected 

Standard Penetration Test (SPT) blowcounts in this sand layer 

have been found to be as low as three in some areas. Because 

the dam may someday experience a peak acceleration as high 

as 0.55 g from a nearby M=6.5 earthquake, it had been concluded that these shallow layers could liquefy and result in a 

failure of the dam. As a result of these evaluations, suspect 

areas were treated in 199 1 by removing and replacing the shallow liquefiable sands downstream of the dam with compacted, 

imported material. Large downstream buttresses were also constructed. 

O'Neill Dam is located about 37 miles from the Loma Prieta 

fault-rupture zone (fig. 1). At the time of the earthquake, the 

reservoir was about 13 feet below the crest of the dam. Strongmotion sensors recorded a peak horizontal acceleration of 0.10 

g at the downstream toe and values of 0.11 g and 0.14 g at two 

locations on the crest. No damage was reported. 

Because of the planned repair to improve seismic stability, it was of interest to compare the predicted versus actual 

performance of the liquefiable sand layers during Loma 

Prieta. To this end the downstream areas of two sites having 

the lowest SPT blowcounts, the station 100 and station 133 

areas, were analyzed to determine the predicted factors of 

safety against the development of liquefaction. The resulting calculated factors of safety against triggering liquefaction during Loma Prieta were between 1.2 and 1.5. These 

"predicted" factors of safety correspond well with the observation of no damage at this dam. 

COSTS OF MAJOR REPAIRS FOR 

EARTHQUAKE DAMAGE 

Although most dams performed reasonably well during 

the earthquake, some required minor to major repairs. The 

approximate repair costs for some of the dams discussed in 

this report are summarized in table 8. 

RESERVOIR ELEVATION ON CREST ELEVATION 234 feet 

OCTOBER 17, 1989 * 221 feet --- 

ZONE 3 BLANKET MODERATELY DENSE 

LIQUEFIABLE SANDS CLAYEY AND SILTY SANDS, 

SILTS, AND CLAYS. 

Figure 17.-Maximum cross section of O'Neill Dam. 
PERFORMANCE OF THE BUILT ENVIRONMENT 

Table 8.-Approximate repair costs for earthquake damage 

Dam Approximate Repair Cost 

Austrian $2,500,000 

Lexington $ 150,000 

Guadalupe $ 100,000 

Chesbro $ 75,000 

Vasona $ 20,000 

' Another $8 million was expended for the design 

and construction of a new spillway as a result of 

the earthquake. 

Table 9.-Peak accelerations measured at earth dams during the Loma Prieta earthquake 

[T, transverse direction; L, longitudinal direction; V denotes vertical direction] 

Maximum Peak accelerations (g) 

Dam height Base Abutment Crest 

(feet) T L V T L V T L V 

Lexington 205 

 an us to 135 .26 -16 .40 .49 .33 

San Luis 313 .04 .06 .02 

.07 .10 .03 

Martinez 54 .09 .08 .02 .14 .15 .04 

Del Valle 222 .04 .06 .03 .08 .08 .07 

Contra Lorna 88 .07 .05 .03 

Note: denotes that other records are available from other instruments at this dam 

STRONG MOTIONS RECORDED ON the studies by Maley and others (1989). Shakal and others 

EMBANKMENT DAMS (1989), and Wood and others (1991). 

Presented in figure 18 is a plot comparing the peak transThe Loma Prieta earthquake provided an excellent oppor- verse accelerations measured at both the base and crest of 

tunity to calibrate dynamic response techniques. As illus- several earth and rockfill dams. These measurements include 

trated in table 9, strong-motion records were recorded at eight those made during the Loma Prieta earthquake as well as 

embankment dams. This information was summarized from those made during previous events. As may be observed, the 
PERFORMANCE OF EARTH DAMS DURING THE LOMA PRIETA EARTHQUAKE D25 

1989 LOMA PRIETA 

EARTHQUAKE 

1 PREVIOUS EARTHQUAKES I 

PEAK TRANSVERSE BASE ACCELERATION (g) 

Figure 18.-Comparison of peak base and crest transverse accelerations measured at earth dams. 

points indicate that at low accelerations, the amplification 

through embankment dams is relatively large. However, as 

the peak base accelerations become larger, the amount of 

amplification is relatively low, possibly a result of increased 

damping or yielding of embankment materials. Also shown 

in figure 18 is a tentative upper bound curve. This curve 

should not necessarily be used for design purposes, but it 

may be useful as a verification tool in the performance of 

dynamic response analyses. 

CONCLUSIONS 

Several earth dams were subjected to relatively strong shaking during the Loma Prieta earthquake and, in general, performed satisfactory. One major dam (Austrian) and one minor dam (Soda Lake) developed moderate damage. A small 

number of other dams developed minor to moderate cracking which required repair. 

A major factor in the good performance of some dams 

was the fact that the reservoirs were commonly at less than 

half their maximum height. Consequently, major portions 

of the embankments were not as saturated as they would 

have been during full reservoir condition. Thus, the earthquake was not the most critical test of these structures. Nevertheless, the performance and strong-motion data will provide researchers with invaluable opportunities to learn more 

about how earth and rockfill dams behave during earthquakes. 

ACKNOWLEDGMENTS 

The purpose of this paper was to present an overview of 

the performance of earth dams during the Loma Prieta earthquake. To this end, the detailed inspection reports prepared 

by the California Division of Safety of Dams following the 

earthquake were invaluable. Providing other details were the 

generally excellent reports prepared by consulting engineers 

that are presented in the list of references. The authors also 

acknowledge Professor Raymond B. Seed and several staff 

members of the California Division of Safety of Dams for 

providing many useful pieces of information. The assistance 

of Thomas Holzer in the publication of this paper is also 

gratefully acknowledged. 

REFERENCES 

Aydin, Atilla, Johnson, Arvid M., and Fleming, Robert W., 1992, 

Right-lateral-reverse surface rupture along the San Andreas 

and Sargent faults associated with the October 17,1989, Loma 

Prieta, California, Earthquake: Geology, v. 20, p. 1063-1067. 

Bureau, Gilles, Babbitt, Donald H., Bischoff, John A., Volpe, Richard L., and Tepel, Robert E., 1989, Effects on dams of the 

Lorna Prieta Earthquake of October 17, 1989: United States 

Committee on Large Dams, Newsletter no. 90, p. 1-4. 

Creegan, Patrick, 1990, Newel1 Dam and the Lorna Prieta earthquake: United States Committee on Large Dams, Newsletter 

no. 91, p. 9-11. 

Lee, Kenneth L., and Roth, Wolfgang, 1977, Seismic stability analysis of Hawkins hydraulic fill: Journal of the Geotechnical Engineering Division, American Society of Civil Engineers, v. 

103, no. GT6, p. 627-644. 

Maley, R., Acosta, A., Ellis, F., Etheredge, E., Foote, L., Johnson, 

D., Porcella, R., Salsrnan, M., and Switzer, J., 1989, U. S. 

Geological Survey Strong-Motion Records from the Northern 

California (Lorna Prieta) earthquake of October 17, 1989: U. 

S. Geological Survey Open-File Report 89-568. 

Rodda, K. V., Harlan, R. D., and Pardini, R. J., 1990, Performance 

of Austrian Dam during the October 17, 1989 Loma Prieta 

earthquake: United States Committee on Large Dams, Newsletter no. 91, p. 7-9. 

Seed, H. Bolton, Makdisi, Faiz I., and De Alba, Pedro, 1978, The 

performance of earth dams during earthquakes: Journal of the 

Geotechnical Engineering Division, American Society of Civil 

Engineers, v. 104, no. GT7, p. 967-994. 

Seed, R. B., Dickenson, S. E., Riemer, M. F., Bray, J. D., Sitar, N., 

Mitchell, J. K., Idriss, I. M., Kayen, R. E., Kropp, A., Harder, 

L. F., and Power, M. S., 1990, Preliminary report on the principal geotechnical aspects of the October 17,1989 Loma Prieta 

earthquake: Earthquake Engineering Research Center, Report 

No. UCBIEERC-90105, University of California, Berkeley. 
D26 PERFORMANCE OF THE BUILT ENVIRONMENT 

Shakal, A., Huang, M., Reichle, M., Ventura, C., Cao, T., Sherburne, 

R., Savage, M., Darragh, R., and Peterson, C., 1989, CSMIP 

strong-motion records from the Santa Cruz Mountains (Loma 

Prieta), California earthquake of 17 October 1989: California 

Division of Mines and Geology, Report No. OSMS 89-06. 

Spudich, Paul, 1996, Synopsis, in The Loma Prieta, California, 

earthquake of October 17,1989-Main-shock Characteristics: 

U. S. Geological Survey Professional Paper 1550-A. 

Volpe, R. L, & Associates, 1990a, Investigation of SCVWD Dams 

affected by the Loma Prieta earthquake of October 17, 1989: 

Report prepared for the Santa Clara Valley Water District. 

Volpe, R. L. & Associates, 1990b, Repair of SCVWD dams affected by the Loma Prieta earthquake of October 17, 1989: 

Report prepared for the Santa Clara Valley Water District. 

Wahler Associates, 1990, Austrian Dam-Investigation and remedial construction following the October 17, 1989 Loma Prieta 

earthquake: Report prepared for the San Jose Water Company. 

Wahler Associates (1992), Seismic Safety Evaluation of Mill Creek 

Dam: Report prepared for Lockheed Missiles and Space Company. 

Wood, Chris, Copeland, David, and Viksne, Andy, 1991, Strongmotion data-Loma Prieta Earthquake of October 17, 1989, 

San Justo dam and dike, San Luis dam, O'Neill dam, Martinez 

dam: U. S. Bureau of Reclamation. 
THE LOMA PRIETA, CALIFORNIA, EARTHQUAKE OF OCTOBER 17,1989: 

PERFORMANCE OF THE BUILT ENVIRONMENT 

EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

ANALYSIS OF SOIL-NAILED EXCAVATIONS STABILITY 

DURING THE 1989 LOMA PRIETA EARTHQUAKE - 

By Mladen Vucetic, University of California, Los Angeles; 

Mark R. Tufenkjian, California State University, Los Angeles; 

Guy Y. Felio, National Research Council, Ottawa, Canada; 

Pirooz Barar, Pirooz Barar & Associates, Oakland, California; and 

K. Ronald Chapman, Schnabel Foundation Co., Walnut Creek, Calif. 

CONTENTS developed and used by Caltrans for calculating the factor of 

safety is suitable for the stability analysis of the grouted soilnailed excavations encompassed by the study. This method 

is based on a bilinear failure surface and the so-called German mode of failure that considers two sliding blocks. 

INTRODUCTION 

Soil nailing is an in-situ technique of mechanically stabilizing soil masses which has been used in Europe for more 

than two decades (Stocker and others, 1979; Chapman and 

Ludwig, 1993; Federal Highway Administration, 1993). In 

North America, as well as in Japan (Japan Highway Public 

Corporation, 1987; Ochiai and others, 1992), soil nailing is 

steadily gaining popularity because it can be used with conventional shoring equipment, it reduces excavation time, it 

allows construction-related activities to proceed in restricted 

space, and it can produce significant savings over conventional shoring techniques in the proper ground conditions. 

The main feature of soil nailing is that it is an in-situ method 

where the existing natural soil is reinforced, as opposed to a 

backfill reinforcement. As shown in figure 1, the inclusions, 

commonly called nails, are installed during the excavation 

using a "top-down" construction procedure, unlike reinforced 

earth walls which are constructed from the bottom up. This 

allows soil retention in areas where little space is available 

for the excavation. The soil-nailing concept is to reinforce 

the soil with passive inclusions, so that the nailed soil mass 

behaves as a composite unit, similar to a gravity retaining 

wall supporting a soil backfill (Juran and Elias, 199 1; Mitchell 

and Villet, 1987). In that sense, soil nailing also differs from 

the conventional tie-back excavation support since the soil 

nails are not prestressed; that is, their resistance can be mobilized only by the movement of soil mass or the face of the 

excavation to which the nails are fixed. Figures 2 to 4 show 

several soil-nailed retaining structures treated in this paper. 

At present, there are three major concerns about soil-nailed 

excavations: (1) the adequacy of the analysis or design methABSTRACT 

The performance of nine different grouted soil-nailed excavations in the San Francisco Bay area during the Loma 

Prieta earthquake is analyzed on the basis of postearthquake 

visual inspections, subsequent stability analyses, and dynamic 

centrifuge model tests. None of the excavations showed any 

signs of movements or similar distress, even though one of 

them was located in the vicinity of the earthquake epicenter 

where there was strong shaking and important seismicrelated damage to other structures. The design and construction practices of grouted soil-nailed excavations in California are discussed. It is concluded that a combination of conservative design and construction is the primary reason for 

excellent seismic stability. It is also confirmed that the method 
D28 EARTH STRUCTURESAND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

ods, (2) the long-term behavior, and (3) the performance 

during seismic loading. Items (1) and (2) have been addressed 

by many researchers, including recently by Gassier (1992), 

Juran and others (1990), Plumelle and others (1990), and 

Stocker and Riedinger (1990). With respect to soil-nailing 

performance during earthquakes, no full-scale field obse~ 

vations were available until the Loma Prieta earthquake. 

During the earthquake, nine soil-nailed structures were subjected to different levels of shaking, including horizontal 

ground-surface accelerations probably as high as 0.4 g. In 

1. Excavation 

10 to 20 degrees WP. 

2. Drilling of nail holes 

also done before the nails 

are installed) 

5. Repeat excavation 

and steps 2 to 4 

Figure 1 .-Steps in the construction of a grouted soil-nailed 

excavation. 

spite of such relatively high horizontal accelerations, these 

structures did not show any visible movements or other signs 

of distress (Felio and others, 1990; Hudson, 1990).A systematic description of these structures and a discussion of 

possible reasons why they performed so well are the main 

purposes of this paper. More details about the corresponding 

investigation can be found inTufenkjian andvucetic (1993). 

SOIL NAILING PRACTICE IN 

CALIFORNIA 

There are three major steps in the construction of a soilnailed wall, as illustrated in figure 1. They are (1) excavation, (2) installation of nails, and (3) construction of facing. 

The excavation generally proceeds in stages ranging from 

1.2 to 1.8 m in depth. One of the major requirements for 

successful soil-nailed systems is that the excavation be capable of self-support for at least a few hours prior to nailing 

and construction of facing. For the most economical construction, however, the self-support should be able to last 1 

to 2 days. As the excavation of each level proceeds, the nails 

are installed at predetermined locations. These reinforcing 

elements may be one of several types: driven, grouted, jetgrouted, or even pneumatically propelled into the ground 

(Myles and Bridle, 1991). However the vast majority of installations are of the open drilled and grouted type (Chapman 

and Ludwig, 1993; Federal HighwayAdministration, 1993). 

In California, and North America in general, the most popular type of nails are the grouted nails, such as those shown in 

figures 2, 3 and 4, since in many locations the soil conditions allow the excavation to stand open long enough. 

Grouted nails generally consist of Grade 60 mild steel bars 

(15 to 45 mm in diameter) placed in boreholes of 100 to 250 

mm diameter. Plastic centralizers are often used to ensure 

proper grout cover of the nail. A cement grout is then placed 

into the boreholes by gravity flow or low pressure.Typica1 

horizontal and vertical spacings range from 1 to 3 m, depending upon the designer's experience and soil conditions. 

The nails are generally inclined at 10' to 20Â from the horizontal. 

Either before or after the nails are in place, a facing structure is built. The facing is required to control soil erosion at 

the excavation face and reduce changes in the moisture content of the soil. The most common type of facing is shotcrete 

layer, 100 to 250 mm thick, which is usually placed by the 

shotcrete method and which is reinforced with welded wire 

mesh. A typical detail of the nail connection to such facing is 

presented in figure 5. If necessary, a blanket of nonwoven 

geotexile is placed between the natural soil and the shotcrete 

to control the drainage. The grouted nail is attached to the 

facing by bolting the steel bar to a square plate usually 300 

to 400 mm wide. For additional reinforcement and strengthening of the facing, horizontal waler bars may be installed to 

connect the plates. Other methods of attachment are used 

for driven nails. For permanent walls, the shotcrete facing 
ANALYSIS OF SOIL-NAILED EXCAVATIONS STABILITY DURING THE 1989 LOMAPRIETA EARTHQUAKE D29 

may not provide for the aesthetic requirements of the project. 

In such cases, either cast-in-place reinforced concrete facing or prefabricated panels can be used. Figure 6 shows photos of large soil-nailed excavation structures recently completed in California. 

THE LOMA PRIETA EARTHQUAKE 

The Loma Prieta earthquake (Ms=7. 1) was one of the most 

costly single natural disasters in U.S. history It caused extensive damage, such as landslides in the epicentral region, 

liquefaction in various areas of the San Francisco Bay region, structural distress to commercial, industrial, and residential buildings, widespread disruption or total destruction 

of utility systems, and damage to critical transportation systems. The earthquake has been the subject of a wide range of 

studies, many of them on geotechnical-related failures, as 

summarized by Seed and others (1991). 

Figure 7 presents an overview of the regional geology and 

the recorded peak horizontal ground-surface accelerations 

during the earthquake. The locations of the nine soil-nailed 

walls considered in this paper are identified on the figure by 

stars, and the location of the epicenter by a circle.The figure 

shows that in the epicentral area the measured maximum 

horizontal ground-surface accelerations, amax, were as high 

as 0.64 g and the vertical up to 0.60g. It can be seen that the 

soil-nailed walls in the northern region (in Richmond, San 

Francisco, Walnut Creek, and San Ramon) were subjected 

to seismic forces corresponding to amax of about 0.10 g. In 

the vicinity of the two walls in Mountain View, an amax of 

around 0.2 g was measured. In the vicinity of the wall in San 

Jose, amax was between 0.11 and 0.1 8 g. The largest amax (0.47 

g) recorded near a soil-nailed wall was in Santa Cruz, some 

16 km due west of the epicenter 

Most of these locations were visited and inspected 2 days 

after the earthquake by a team from the University of California, Los Angeles (Felio and others, 1990), and some walls 

were inspected subsequently by design and construction companies. As stated earlier, no signs of distress or corresponding deformation were found on the walls, indicating excellent performance of such structures during moderate and 

strong shaking. 

1.1 m min. 

existing grade 

shotcrete facing 

sub 

ele. 

Figure 2.-Cross section of a soil-nailed excavation for a building constructed in Santa Cruz (UCSC wall), Calif. 

(Felio and others, 1990). 
D30 EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

CHARACTERISTICS AND DESCRIPTION 

OF THE WALLS 

The main characteristics of the nine walls inspected after 

the earthquake are summarized in table 1. In figure 8 the 

dimensions of the walls are presented in a uniform scale. 

Thevariation of the geometry, characteristics of the walls, 

soil conditions, and estimated ground-surface accelerations 

are evident. The walls are further characterized in table 2 in 

terms of the following three dimensionless ratios commonly 

used as design criteria (Bruce and Jewell, 1987): 

maximum nail length 

Length ratio = (1 

excavation height 

hole diameter x nail length 

Bond ratio = (2) 

(nail 

Strength ratio = (3) 

horizontal spacing x vertical spacing 

Table 3 further compares the three dimensionless ratios 

computed for the nine San Francisco walls, with the values 

computed by Bruce and Jewell (1986, 1987) for soil-nailed 

structures with drilled and grouted nails constructed all over 

the world. The bond and strength ratios generally fall within 

the range of other soil-nailed retaining structures. However, 

the length ratios for the San Francisco walls are generally 

much higher than those calculated from other sites, suggesting that the San Francisco walls are more conservatively 

designed. Note from table 2 that the length ratio for the UC 

Santa Cruz wall is the smallest. This wall is apparently the 

least conservatively designed of all of the walls, yet it is located in the vicinity of the highest estimated peak horizontal 

ground-surface acceleration. In spite of these facts, no obhorizontal spacing x vertical spacing 

i!L-L&/tt=,,, # SOIL NAIL X 6.1 m LONG 

WELDED WIRE MESH 

/lh-,,.-- 190 mm DIA. DRILLED HOLE 

MIRA 

Y I 0.4 m X u J m DEEP FOOTING 

WITH CONCRETE 

0.3 m WIDE X 0.3 m 

DEEP FOOTING 

Figure 3.Ã‘Cros section of a permanent soil-nailed retaining structure located in San Ramon (NME wall), Calif. (Barar, 1990). 
-- 

ANALYSIS OF SOL-NAILED EXCAVATIONS STABILITY DURING 'THE 1989 LOMA PRIETA EARTHQUAKE D3 1 

servable damage was noted on the Santa Cruz structure after 

the earthquake. The walls are described in greater detail below. 

MOUNTAIN VIEW, 2350 EL CAMINO REAL (ECR 

WALL) 

Nearly 280 m2 of soil-nailing construction was used to 

provide temporary shoring of an excavation for an office 

building. The concrete wall for the new structure was to be 

poured in front of the soil-nailed concrete facing. The subsurface soil consisted of gravelly and clayey sand. The shearstrength parameters used in design were c = 9.6 kN/m2 and f 

= 30'. while the soil unit weight was assumed to be 17.3 kN/ 

m3. The soil-nailed wall was completed by May of 1989 and 

the excavation was still open when the earthquake struck. 

Postearthquake observations revealed only a few shallow 

hairline cracks in the concrete facing, typical of flexural 

cracking if the facing is considered as a vertical slab with 

the nails acting as reaction points. Note in table 1 that the 

facing of this wall was relatively thin (100 mm), while the 

estimated horizontal acceleration was considerable (0.2 1 to 

0.27 g). 

MOUNTAIN VIEW, KAISER PERMANENTE 

PARKING GARAGE (KPG WALL) 

Approximately 380 m2 of shoring was provided for the 

construction of a parking garage. Soil nailing was used only 

on one side of the excavation, while a combination of other 

shoring techniques were used on the remaining sides. The 

soil conditions at the site consisted of stiff sandy to clayey 

2nd floor 0 -- - 

1st floor 

300 mrn diameter nails 

4- 6.1 to 8.5 m long 

footing 

Figure 4.Ã‘Cros section of a soil-nailed retaining wall constructed on a steep hill in San Francisco (CVA wall), Calif., 

to provide adequate space for an apartment complex (Felio and others, 1990). 
D32 EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

silt overlying silty to sandy clay. The shear strength param- struction of the shoring was completed just 8 days before the 

eters used in design were c = 23.9 kN/m2 and f = 14O, while earthquake. The postearthquake observations revealed no 

the soil unit weight was assumed to be 18.8 kN/m3. The con- visible distress to the soil-nailed wall, while the opposite side 

Shotcrete 

t-eÃ‘Ã‘Ã‘6Ã‘ 

First application 50 mrn Second application 50 mrn 

100 mm dia. bore hole filled with grout 1 4-#4 horizontal waler bars 

#8 Rebar (typical) 

150 rnrn X 150 mm X 13 mm plate 

100 rnm X 100 mm - 818 mm wire mesh 

Figure 5.Ã‘Connectio between grouted nail and facing. A, Cross section of a typical connection (from Koerner, 1984). 

B, Strong reinforcement around the nail tip. 
ANALYSIS OF SOIL-NAILED EXCAVATIONS STABILITY DURING THE 1989 LOMA PRIETA EARTHQUAKE D33 

Figure 6.-Soil-nailed structures recently completed in California. A, Soil-nailed excavation for an underground structure of a building. B, A highway retaining soil-nailed wall showing different stages of the construction of permanent 

facing. 
D34 EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

Figure 7.-Overview of regional geology and recorded peak horizontal ground-surface accelerations during the Lorna Prieta earthquake 

(from Seed and others, 1991). 
ANALYSIS OF SOIL-NAILED EXCAVATIONS STABILITY DURING THE 1989 LOMAPRIETA EARTHQUAKE D35 

of the excavation, which used cantilevered soldier beams 

with a concrete facing between the beams, revealed some 

vertical hairline cracks in the facing. 

SANTA CRUZ, UNIVERSITY OF CALIFORNIA AT 

SANTA CRUZ (UCSC WALL) 

Approximately 350 m2 of shoring was required to construct a new science library on the UCSC campus.The soil 

conditions at the site consisted of sandy silt to sandy clayey 

silt extending from the ground surface to a depth of approximately 6.4 m. The soil has an average dry unit weight of 

13.8 kN/m3 and a moisture content ranging from 26.2 percent in the clayey silt near the surface to 13.9 percent in the 

sandy silt at 6.4 m. Shear strength properties used in design 

were c = 23.9 kN/m2 and f = 25'. The cross section at the 

highest location of the wall is shown in figure 2. Construction of the wall was completed on September 28, 1989, less 

than 3 weeks before the earthquake. Since three sides of the 

excavation were soil nailed, at least one side may have been 

subjected to the full strength of the earthquake, which pro1. Mountain View ECR 6. San Ramon m~ 

2. Mountain View KPG 

3. Sam cm UCSC 

4. San Jose RPP 

5, San Jose RPP 

7. San Francisco CVA 

9. Richmond TSW 

Figure 8.-Dimensions of the investigated soil-nailed walls, 

duced in the vicinity of the excavation peak horizontal 

ground-surface accelerations of about 0.47 g. It should be 

noted that this wall was located closest to the epicenter and 

presumably was subjected to the strongest shaking, while at 

the same time it had the smallest ength ratio and thinnest 1 facing among the nine walls examined (see tables 1 and 2). 

Prior to the earthquake, some wall and column spread footings had been poured (see fig. 2). A postearthquake inspection revealed significant cracking in the concrete of the footings. This cracking was not attributed to shrinkage since foundations constructed after October 17 showed fewer cracks. 

As opposed to that, the inspection 4f the soil-nailed wall after the earthquake revealed no cracking. A week after the 

earthquake, nine nails were tested to 150 percent of their 

design pull-out load. The tests showed no loss in the carrying capacity of the nails due to the seismic activity 

SAN JOSE, RIVERPARK PROJECT (2 RPP WALLS) 

These two retaining walls were designed and built as peF 

manent structures along the Guadalupe River in San Jose, 

approximately 40 km north of the epicenter The subsurface 

soil consists of silty and sandy clays to a depth of about 4.5 

to 6 m. According to the geotechnical report, these clays have 

an intermediate to high plasticity with an approximate aveF 

age dry unit weight and moisture content of 14.1 kN/d and 

22 perecent, respectively, and an undrained shear strength 

ranging from 72 to 240 kN/d, as interpolated from static 

cone penetration tests. The clays are underlain by a 3 m zone 

of dense, clayey, silty, gravelly sand with an average dry 

unit weight and moisture content of about 17.3 kN/d and 

15 percent, respectively. The shear strength parameters used 

in design were c = 23.9 kN/m2 and f = 0Â° while the total unit 

weight was assumed to be 19.6 ksld. Since these are permanent walls, the concrete surface was finished off with architectural concrete and in some places clad with granite. 

The postearthquake observations ievealed no signs of distress. 

SAN RAMON, NATIONAL MEDICAL ENTERPRISES COMMUNITY HOSPITAL (NME WALL) 

This soil-nailed retaining structure forms a part of a peF 

manent retaining wall used for the roads and landscape that 

surround the medical centeK The wall cross section is shown 

in figure 3. According to the geotechnical report, the soil 

conditions consist mainly of engineered fill up to a maximum depth of 24 m, generated from cut-and-fill operations 

performed previously. Therefore, the soil-nailed retaining 

structure was built in fill material. The fill consists of sandy 

and silty clay of moderate to high plasticity with an average 

dry unit weight and moisture content of approximately 17.1 

kN/m3 and 18 percent, respectively The shear strength prop- 
D36 EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

Table 1 .-Summary of soil-nailed walls investigated 

Project hation Height Nail details Facing General Soil Shear strength Estimated 

No. (see fig. 7) of wall Spacing Length Diameter Inclination thickness soil unit design parameten horizontaJ 

(m) (m) (m) &out Rebx (m) weight c 4' ground 

(mn.0 (m) Typ (m/m3 (m,m2) (cie.gms) Mace 

1 

acceleration (R) 

Mountain 

View 

ECR 

Mountain 

View 

KKi 

Smta 

cmz 

ucsc 

San Jose 

RPP 1 

San Jose 

RPP2 

San Ramon 

NhE 

San 

Francisco 

CVA 

Walnut 

ckek 

MSW 

Richmond 

1.7V 

1.5H 

2.0v 

1.5H 

1.5V 

1.8H 

1.8V 

1.8H 

1.8V 

1.8H 

1.8V 

1.8H 

1.8V 

1.5H 

(one 

mw) 

1.8H 

1.8V 

clayey 

sand 

sandy to 

clayey silt 

sandy to 

clayey silt 

alluvial 

clay, silt 

andsand 

alluvial 

clay, silt 

and sand 

engineered 

fill 

fill over 

silty clay 

and highly 

weatheld 

siltstone 

fill over 

medium 

stiff to stiff 

clay 

alluvium 

TS W 1.8H 9.1 190 25 15 200 deposits 18.8 19.1 28 0.05 to 0.15 

erties used in design were c = 47.9 kN/m2 and f = OO. Since SAN FRANCISCO, CRESTA VISTA APARTMENTS 

this is a permanent structure, the facing of the soil-nailed (CVA WALL) 

wall was finished with a colored architectural concrete finish. The postearthquake walk-through revealed that the sur- This wall demonstrates the unique concept of using soil 

face of the concrete remained smooth and free of cracks. nailing on a permanent basis to retain the slope and cut on a 

Table 2.-Dimensionless ratios for Sun Francisco area soil-nailed walls 

Length ratio Bond ratio Strength ratio (lo3) 

Project No. Location max. nail leng th hole dia. x nail le ngth (nail dia.)2 

excav. height H . spacing x V. spacing H . spacing x V. spacing 

1 Mountain View 0.85 0.6 1 0.40 

I 

ECR 

2 Mountain View 0.85 0.34 0.2 1 

KPG 

3 Santa Cruz 0.74 0.38 0.38 

ucsc 

4 San Jose 1.6 0.34 0.19 

RPP 

5 San Jose 2.5 0.34 0.19 

RPP 

6 San Ramon 1.1 0.36 0.19 

NME 

7 Sm Francisco 1.5 to 1.7 0.68 to 0.94 0.28 

CVA 

8 Walnut Creek 1.7 not applicable not applicable 

MSW 

9 Richmond 1.2 0.36 to 0.53 0.19 

TSW 
ANALYSIS OF SOIL-NAILED EXCAVATIONS STABILITY DURING THE 1989 LOMAPRIETA EARTHQUAKE D37 

Table 3.-Dimensionless ratios for soil-nailed walls 

Drilled and grouted in Drilled and grouted in 

granular soils morraine and marl San Francisco 

(Bruce and Jewell, (Bruce and Jewell, walls 

1986, 1987) 1986, 1987) 

Length ratio 0.5 to 0.8 0.5 to 1.0 0.7 to 0.25 

Bond ratio 0.3 to 0.8 0.15 to 0.20 0.34 to 0.94 

Strength ratio 0.4 to 0.8 0.1 to 0.25 0.19 to 0.40 

steep hill to make room for the development of a housing 

project. The wall cross section is shown in figure 4.The 9.8- 

m-high soil-nailed structure was constructed at the toe of a 

45.7-m-high slope to allow for the construction of apartment 

units. The wall is about 90 m long and consists of two levels. 

Due to the permanent nature of the structure, a 200 mm reinforced concrete facing and a small footing at the base were 

used. The soil conditions at the site can be described as colluvium and residual soil deposits.The design parameters used 

were cohesion c = 9.6 kN/m2 and an angle of internal friction off = 35'. The inspection that took place 3 days after 

the earthquake showed no signs of distress to the wall and 

no indications of lateral movements or tension cracks in the 

hill behind the wall. 

WALNUT CREEK, MINI STORAGE FACILITY 

(MSW WALL) 

The project consisted of a three-story building with two 

levels above grade and one level of basement below The 

soil-nailed wall was integrated into the final basement wall. 

The soil at the site consists mainly of fill material up to 3 m 

depth, including a nonuniform mixture of gravel, sand, and 

clay. The underlying soil consists of stiff silty clay. The average dry unit weight and moisture content of the fill is 16.5 

kN/m3 and 20 percent, respectively The shear strength parameters assumed in design were c = 14.4 kN/m2 and f = 

28', while the total unit weight of the soil was assumed to be 

18.8 kN/m3. The postearthquake observations revealed no 

signs of distress on the surface of the wall or at grade behind 

the wall. 

RICHMOND, TEMPORARY SHORING WALL 

(TSW WALL) 

Soil nailing was used here to construct a temporary sho~ 

ing wall which has the tallest single-level vertical face of 

any of the walls examined in this papex A permanent retaining wall was eventually built in front of the soil-nailed wall. 

Unfortunately, soil stratigraphy data is not available for this 

site. However, the shear strength parameters used in design 

were c = 19.1 kN/m2 and f = 28', while the unit weight of the 

soil was assumed to be 18.8 kN/d. A walk-through of the 

site following the earthquake did not reveal any signs of distress attributable to seismic activity. 

METHODS OF' ANALYSIS 

Most of the current design methods for soil-nailed retaining structures under static loads are derived from classical 

slope-stability analyses, which incorporate a limit equilibrium approach. Accordingly, they evaluate global factors of 

safety along assumed failure surfaces such as those shown 

in figure 9. They are usually referred to as the German method 

(Stocker and others, 1979; Gassler and Gudehus, 1981; 

Lambe and Jayaratne, 1987), Davis method (Shen and others, 1981; Bang and others, 1992), French method (Schlosser 

and others, 19831, and Caltrans method (computer program 

SNAIL: Caltrans, 1993). The differences in the methods result from the definition of the factor of safety assumed failure surface shape, and the assumed contribution of the soil 

nails to the stability. In that respect, the methods are contradictory, and because of the lack of full-scale observations of 

actual failure mechanisms, different points of view about their 

applicability have emeqed. 

The German method (fig. 9A) assumes a bilinear failure 

surface passing through the toe of the excavation.The failing soil mass is broken into two parts.The first part contains 

most of the nailed soil mass, while the second part forms the 

active earth pressure wedge behind it-behind the "soilnailed gravity wall." The analysis considers the tensile and 

pull-out resistance of the nails crossing the failure surface 

and, of course, the forces of interaction between the nailed 

mass and active wedge behind it. The assumed failure surfaces are consistent with the concept of soil nailing, that is, 

the nailed soil mass behaves like a reinforced block. 

The Davis method incorporates a parabolic failure surface 

that also passes through the toe, as shown in figure B. The 

sliding surface either passes entirely through the nails or intersects the ground surface somewhere beyond the reinforced 

zone. In the analysis, the tensile and pull-out resistance of 

the nails crossing the failure surface are considered the governing stabilizing forces. Because of its successful track 
D38 EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

record and easy implementation, it has been a popular design method in the United States. This has been the case in 

spite of the fact that the assumption of a parabolic slip surface (which does not change slope when crossing from the 

nonreinforced to the reinforced zone) has not been adequately 

verified by laboratory or field tests. 

The French method follows procedures similar to the Davis 

method, but assumes a circular failure surface passing entirely through the nails, as shown in figure 9C. But, unlike 

the previous two methods, this method considers the shear 

and bending of the nails, which adds to the complexity of 

the analysis. 

The Caltrans method also assumes a bilinear failure surface, just like the German method. However, unlike in the 

German method, the bilinear failure surface may pass entirely through the nails (see fig. 9D). 

More recently, a kinematical limit analysis approach has 

been proposed for the design of soil-nailed retaining structures (Juran and others, 1990). It differs from the other analysis procedures in that it suggests a method for estimating 

nail forces. In this way, it may provide a check on local stability at each level of nail reinforcement. The method assumes that the failure surface is defined by a log-spiral passing partially through the nails and that the failure occurs by 

rotation of a quasi-rigid body along this surface. 

All of the San Francisco walls examined in this study were 

designed using a modified version of the Davis method 

(Barar, 1990; Felio and others, 1990). Seismic forces were 

accounted for by using an equivalent static horizontal force 

H = W x kh, applied at the center of gravity of the potentially 

unstable soil nailed mass, where W is the weight of the moving soil mass and kh is the horizontal seismic coefficient. 

It should be mentioned at this point that the Davis method, 

as well as the German and Caltrans methods, has a certain 

degree of the inherent conservatism in that the potential stabilizing effects of the shear and bending resistances of the nails 

are ignored. New studies (Jewell and Pedley, 1992; Federal 

Highway Administration, 1993) show, however, that the effects of bending stiffness are small. Also, the contribution of 

the steel reinforced facing to the strength of the system is 

unaccounted for. The lack of full understanding of the role 

of facing in the global and local stability apparently led to 

the difference by a factor of 3 (75 mm vs. 200 mm) in the 

thicknesses of the facing among the nine walls under consideration. Some designers and construction companies feel 

comfortable 'with thinner facing, while some prefer more 

conservative thicker facing. Figure 5, for example, illustrates 

a rather heavily reinforced facing with a sturdy nail contact. 

The role of the facing in soil reinforcing stability is just beginning to be studied as a separate issue (Tatsuoka, 1992), 

and it should definitely be given more attention in the future. 

The factors of safety for the CVA and TSW soil-nailed 

walls obtained by the Davis method, modified to account 

for earthquake forces by the pseudostatic technique, are presented in table 4. The location of the assumed failure surfaces that yield minimum factors of safety for the TSW wall 

are shown in figure 10. In general, the factors of safety are 

relatively low, especially for the range of estimated peak 

horizontal ground accelerations during the Loma Prieta earthquake. According to such low factors of safety, and given 

the fact that some soil-nailed structures were probably subjected to much larger horizontal forces, some visible damage should have occurred during the earthquake. This should 

have been expected in particular for the UCSC Wall in Santa 

Cruz, which had the smallest length ratio and facing thickness, and yet is likely to have undergone horizontal seismic 

forces as large as 0.4 g. The lack of visible damage on any of 

the walls, except very thin cracks on the ECR wall, suggests 

that either the design, analysis, or construction, or most likely 

their combination, may have been more conservative than 

necessary. The lack of damage also indicates that the assumed 

failure surface and mechanism of failure of the Davis method 

may not be fully appropriate for the nine walls treated here. 

In the following section, the components of the analysis, 

design, and construction that appear to be on the conservative side, and therefore could be responsible for such excellent seismic performance, are discussed. 

POSSIBLE IWASONS FOR THE 

OBSERVED BEHAVIOR 

Since soil nailing is a relatively new soil-stabilization technique, with very little practical experience of full-scale static 

failures and practically no experience of seismic failures, 

the design and construction are usually quite conservative. 

The preliminary design of a soil-nailed retaining structure 

proceeds much like that of retaining walls, by trial and error. 

Based mainly on the expected excavation height and the soil 

strength properties, tentative characteristics of nails and facing (length, diameter, horizontal, and vertical spacings of 

nails, and the thickness and reinforcement of facing, etc.) 

can be assumed and some sort of stability analysis performed. 

The assumed values and characteristics depend primarily on 

the designer's experience with other satisfactorily constructed 

soil-nailed walls, which may lead to an overly conservative 

design, and to a lesser extent on charts and dimensionless 

parameters derived by others, such as those by Bruce and 

Jewell (1986,1987) and Guilloux and Schlosser (1982). Table 

3 shows, for example, that the length ratios and bond ratios 

for the nine walls considered here are on the conservative 

side in comparison with the values suggested by Bruce and 

Jewell (1986, 1987). 

The main components of the conservative design and construction for seismic loads include (1) conservative and most 

probably unrealistic assumption of the failure mechanism, 

(2) no consideration of the contribution of the facing in the 

stability analysis, and (3) conservative construction due to 

the lack of field experience and understanding of the various 
ANALYSIS OF SOIL-NAILED EXCAVATIONS STABILITY DURING THE 1989 LOMA PRIETA EARTHQUAKE D39 

aspects of soil-nailed excavation seismic response. The first 

two components are discussed below. 

F'AILUW-MECHANISM ASSUMPTION 

Due to a lack of full-scale observations of failures and 

corresponding failure mechanisms under both static and seismic loads, there is currently no consensus among designers 

on which failure mode is the most realistic among the four 

basic modes presented in figure 9. To cast more light on possible modes of failures under dynamic loads, two series of 

Circular 

Figure 9.-Assumed failure surfaces used in analyses. A, German 

Method. B, Davis Method. C, French Method. D, Caltrans Method. 

dynamic centrifuge tests were conducted, one in 1991 

(Tufenkjian and others, 1991; Tufenkjian and Vucetic, 1992; 

Vucetic and others, 1993) and the other in 1996 (Vucetic 

and others, 1996). Figures 11 and 12 show the main features 

of the models tested and results obtained in 1991. 

The centrifuge tests were performed at the Rensselaer 

Polytechnic Institute (RPI) Geotechnical Centrifuge Research 

Center on a 3-m radius Accutronic 665- 1 centrifuge (Elgamal 

and others, 1991). The scale factor was 50 in all of the tests. 

Accordingly, to simulate prototype geostatic stresses, the 

models had to undergo a centrifugal acceleration of 50 g. 

For dynamic testing, a servo-hydraulic earthquake simulation shaker mounted on the centrifuge platform was used. 

Four models were tested in 199 1. 

They represented 7.6-m-high soil-nailed excavations with 

grouted nails, corresponding roughly to an excavation height 

of a two- to three-story underground garage. The effects of 

two important characteristics of soil-nailed structures were 

tested: the length of nails (expressed in terms of the length 

ratio), and the axial and flexural rigidities of the nails. 

Three length ratios were tested, 0.33,0.67, and 1.0, which 

could be characterized as the ratios corresponding to short, 

medium, and long nails (see table 3). These ratios cover approximately five out of the nine walls listed in table 2. The 

four other walls have very large length ratios between 1.5 

and 2.5. Two axial and flexural rigidities of the nails were 

used, one that can be considered regular and the other than 

can be considered small. By varying the axial and flexural 

rigidities of the soil nails, their effect on the failure surface 

geometry and stability could be assessed. As shown in figures 11A and 11 C, three displacement transducers (LVDT9s) 

were used to record the lateral movements of the facing and 

the vertical soil settlement behind the facing. During dynamic 

loading, four accelerometers were utilized to measure the 

accelerations of the model box and in various locations within 

the model box. The soil used in the experiments was fine 

sand. The sand was partially saturated to generate an apparent cohesion, necessary for a rough simulation of in-situ cohesion and cementation. Other details of the 199 1 testing 

are described by Vucetic and others (1993). 

Figures 11B and 12 show a typical failure mechanism obtained in the tests under horizontal dynamic loads. In all four 

tests the failure surface never started at the ground surface 

above the nails. Instead, it started at the ground surface behind the ends of the nails. Figure ll B reveals that the failure 

mechanism involves three soil "zones" and two soil ''blocks," 

with two failure surfaces, one of which consists of two parts. 

The primary failure surface extends from behind the nails at 

the ground surface down to the end of the second row of 

nails, at which point it changes curvature and continues down 

to the bottom of the excavation through the toe. The secondary failure surface develops within the sliding soil mass and 

divides zones 1 and 3. Such deformation patterns after the 

tests point to the following failure mechanism. The soil above 

the second row of nails in zone 1 moves horizontally under 
EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

Table 4.-Calculated factors of safety using the Davis method (see also Hudson, 1990) 

Horizontal acceleration Cresta Vista apartments Temporary shoring wall 

coefficient, k,, CVA TSW 

0 1.31 1.19 

Indicates the factors of safety corresponding to the range of estimated horizontal peak 

ground-surface accelerations near the site. 

large inertial forces as a relatively rigid block held together 

by the nails. Consequently, the soil in zone 2 is pushed outward by the horizontal friction along the interface between 

the upper zone 1 and the lower zone 2.Accordingly, the failure surface passes through the bottom row of nails. In such a 

mechanism, the bottom nails obviously act as anchors between the back soil and the facing, while the top nails hold 

the soil together in the upper part of the excavationAs zones 

1 and 2 move horizontally outward during seismic shaking, 

the lateral stresses in zone 3 are greatly reduced. Consequently, zone 3 represents a typical failure wedge behind a 

retaining wall, the retaining wall being zone 1.This mechanism and kinematics of the soil movement resemble the geometry of German method for static stability evaluation, 

shown in figure 9A (Gassler and Gudehus, 198 I), while they 

contradict the assumption that rotation of one monolith occurs along a continuous circular or parabolic failure surface. 

To examine the factors governing the failure corresponding 

to the above mechanism, the forces and factors of safety for 

the TSW Wall in Richmond (see fig. 10 and table 4) are 

reevaluated. Figure 13 shows the assumed failure surfaces 

and governing forces, while figure 14 shows the 

corresponding polygons of forces. To account for the effects 

of dynamic horizontal forces the pseudostatic method of 

analysis is used again, where the dynamic action is 

represented by the static horizontal force H = W x kh. Two 

definitions of the factor of safety, FS, based on the German 

Design parameters: 

7 = 18.8 kN/m3 

c = 19.1 kN/m2 

"zone of assumed parabolic failure 

surfaces passing through the nails. 

Figure 10.-Assumed failure surfaces passing through the nails for 

the factor of safety evaluation of theTSW wall in Richmond using 

Davis method and its modifications. 

type of failure mechanism are considered below First is the 

definition for static stability proposed by Stocker, Korber, 

Gassler and Gudehus (1979), which is adapted here for the 

dynamic stability by adding to the poligon of forces a 

horizontal force H = W x kh. The second is the definition 

proposed and used by Caltrans (1993).Accordingly, the two 

methods for the calculation of FS are called here the SKGG 

method and the Caltrans method. 

According to the SKGG method, the factor of safety is 

calculated as 

where Za 

beyond the 

= cumulative axial pull-out force of the nails 

failure surface and Ze = mobilized cumulative 

axial force of the nails beyond the failure surface. Therefore, the entire factor of safety is based on the pullout of the 

nails. For the TSW wall, FS for different seismic coefficients, kh, was calculated. The results of this calculation are 

presented in terms of the FS vs. kh relationship in figure 15. 

On the same figure the equivalent relationship betweenFS 

and kh obtained by the Caltrans method is presented as well. 

The bilinear failure surface assumed in the Caltrans 

method is similar to the failure surfaces assumed in the SKGG 

method and thus to the deformation patterns and failure surfaces observed in the centrifuge tests. In fact, as indicated 

in figure 13, the forces and their positions relative to the free 

bodies for the TSW wall are the same for the SKGG and 

Caltrans methods. However, the methods differ fundamentally in their definitions of the factor of safety. The Caltrans 

method applies a unique factor of safety to the soil cohesion, 

c, soil friction angle, 0, and the cumulative nail pullout force, 

za : 

c' = c/FS = mobilized cohesion, 

0' = tad [(tan 0)1FS\ = mobilized friction angle, and 

Ze = Za/FS = mobilized pullout force. 

The method then utilizes these "mobilized" parameters in 

the force equilibrium equations to solve for the interwedge 

forces, F (see figs. 13 and 14). Since these forces must be 

equal in magnitude and opposite in direction, assumed value 

of FS is systematically varied until this condition is fulfilled, 

which then yields the corresponding FS used in design. The 

Caltrans method has been coded into a computer program- 
ANALYSIS OF SOIL-NAILED EXCAVATIONS STABILITY DURING THE 1989 LOMA PRIETA EARTHQUAKE D41 

ming language and can be run on personal computers. The 

computer program is called SNAIL (Caltrans, 1993). By inAccelerometer 

LVDT 

TIME (ms) c r 

Figure 11.-Features of the typical centrifuge model test with 

length ratio of 0.67 (Tufenkjian and others, 1991; Vucetic and others, 1993). A, Longitudinal cross section of the soil-nailed excavation centrifuge model box. B, Failure mechanism obtained in the 

centrifuge due to strong horizontal shaking. C, Typical records of 

soil mass movements during shaking with 10 cycles of 0.27 g cyclic acceleration amplitude. 

putting the geometry of the slope and details of the soil 

strength and nail properties, the program can systematically 

vary the location of the bilinear failure surfaces, until the 

one producing the lowest factor of safety is found. The program also has an option to calculate the FS for a specified 

surface, as well as for considering seismic forces by the 

pseudostatic technique. 

Several interesting conclusions can be derived from figure 15. First is that for FS =1 (the conditions of the failure of 

the wall), kh "0.37 is obtained by both methods. Second, 

this k value is much larger than k, 4.1 to 0.20 corresponding to FS = 1 calculated according to the Davis method (see 

table 4). Third, kh ~0.37 isin relatively good agreement with 

the amplitude of the cyclic acceleration of 0.45 g that was 

required in the centrifuge testing for the failure of the soil-11 

nailed excavation model of similar length ratio (see Vucetic 

and others, 1993). And fourth, the FS versus kh relationship 

for the SKGG method has a singularity point, while the same 

relationship for the Caltrans method does not. 

Based on these observations it can be concluded that the 

failure mechanism according to the German method seems 

to be more appropriate than that of the Davis method. 

However, figure 15 also shows that using the SKGG method 

to calculate the factor of safety may not be suitable for 

the calculation of stability involving the horizontal forces 

(W x k ), because it is too sensitive to the variation of k,, . By varying kh from 0.3 to 0.4, FS varies from -1 to -oo and 

then from +OO to 0.7-that is, as noted above, the function 

FS = f (kh) has a singularity point. The reason for such 

sensitivity of FS with respect to kh can be easily understood 

from the polygon of forces in figure 14B. For example, if the 

force H1=Wlx k is increased by only 15 percent, the Z force 

will double-that is, change by 100 percent. Consequently, 

the FS = Z/Z will change dramatically too. Such sensitivity 

of FS comes from the fact that FS is defined on the basis of 

forces which are of secondary importance for the stability of 

the structure. In other words, force Z is relatively small 

compared to the other forces in the polygon. More dominant 

forces are apparently the reaction force Q and cohesion force 

C mobilized along the failure surface. 

In the configuration of forces such as shown in the polygon in figure 14B, corresponding to the German method type 

of failure mechanism, the role of the nails is predominantly 

to interact with the soil and form the nailed block. Such a 

large soil block is evidently seismically very stable, and its 

stability is governed by the large forces of friction and cohesion at the interfaces with the surrounding soil, not by the 

small forces such as Z This, of course, would change if Z is 

relatively large, that is, corresponding to very long nails installed deep beyond the failure surface. In such a case the 

kinematics of the failure would be different too. Instead of 

predominantly sliding along the failure surfaces, the facing 

and thus the soil mass would be forced to rotate around the 

bottom row of nails which are anchored beyond the failure 

surface. 
EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

Figure 12.-Excavated model of a test with the length ratio of 0.33 (short nails) shaken by 10 cycles of the uniform 

acceleration amplitude of 0.10 g. 

As opposed to the SKGG method of stability evaluation, 

the Caltrans method does incorporate all significant forces 

in the definition of FS and consequently yields a more meaningful relationship between FS and k,. This, along with the 

above discussion, leads to the following conclusions: (1) the 

German-method type of failure mechanism seems to be appropriate, (2) the Caltrans method for calculating FS seems 

to yield appropriate and meaningful results, and (3) the Davis 

method used in the design of the San Francisco Bay area 

soil-nailed walls seems to be overly conservative, apparently 

because it employs an unlikely failure mechanism. 

To confirm the above conclusions, the factors of safety of 

the eight soil-nailed walls subjected to the Loma Prieta earthquake were calculated by the Caltrans method. The fourlevel RPP wall located in San Jose could not be accurately 

reproduced using the SNAIL program. The wall geometries 

were scaled from figure 8 and the soil design parameters 

were taken from table 1. The calculated factors of safety are 

shown in table 5, where those corresponding to the range of 

estimated horizontal ground-surface accelerations near each 

soil-nailed wall are indicated by double-headed arrows. Note 

that the factors of safety are generally much greater than unity 

for the range of estimated ground-surface accelerations, even 

for the UCSC wall which was subjected to horizontal accelerations in the range of 0.4 to 0.5 g. Recall from Table 4, for 

example, that the Davis method predicted failure (FS = 1 .O) 

for the TSW wall for the acceleration coefficient between 

Figure 13.-Failure surfaces and forces for the TSW wall employed 

in both SKGG and CALTRANS methods for the calculation of the 

factor of safety according to the German type of failure. 
ANALYSIS OF SOIL-NAILED EXCAVATIONS STABILITY DURING THE 1989 LOMA PRIETA EARTHQUAKE D43 

0.10 and 0.20, while the Caltrans method required an acceleration coefficient between 0.4 and 0.5. Such large values of FS obtained by Caltrans method are in agreement with 

the excellent performance of the walls during the earthquake. 

ROLE OF FACING 

None of the methods discussed above account explicitly 

for the contribution of the facing in the evaluation of the 

global factor of safety, although they do incorporate the evaluation of punching shear around the nail connection. In other 

Block: ZONE 3 

Block: ZONES AND 2 

B 

Figure 14.-Polygons of forces identified in figure 13. A, Polygon 

for block comprising zone 3. B, Polygon for block comprising zones 

1 and 2. 

words, the factor of safety is calculated without considering 

axial and flexural rigidities of the facing. In that respect, there 

is no consensus on what the contribution of the facing to the 

global stability of soil nailed structure really is. However, it 

is obvious that stronger facing and stronger contact between 

the facing and the nails will make the nailed soil mass more 

coherent. The failure mechanism of such a coherent soil mass 

is likely to be of the German type-that is, behaving as a 

large seismically stable block. In addition, the inability of 

the nails (which are firmly fixed to the facing) to move freely 

decreases the likelihood of local failures, especially in the 

zones most critically stressed during construction and seismic loading. As suggested earlier, the lack of full understanding of the role of facing in the global and local stability apparently led to the difference by a factor of 3 (75 rnm vs. 200 

mm) in the thicknesses of the facing between the nine walls 

considered here. 

CONCLUSIONS 

Postearthquake inspections of nine soil-nailed walls following the Loma Prieta earthquake indicated superior performance and no signs of distress, even though one of the 

walls was subjected to horizontal accelerations probably as 

high as 0.4 g. It was shown that the excellent performance 

may be attributed to a conservative design, generally conservative stability analysis which is mainly the result of an 

Figure 15.-Variation of the factors of safety, FS, with the seismic 

coefficient, kh, for the TSW wall. 
D44 EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

Table 5.-Calculated factors of safety using the Caltrans method for San Francisco area soil-nailed walls 

Horizontal Soil-nailed wall 

acceleration ECR KPG UCSC RPP NME CVA MSW TSW 

coefficient, k,, (singlelevel) 

0.0 2.28 2.53 3.30 1.59 1.64 

0.1 2 .OO 2.26 

0.2 1.27 c 1.35 1.17 c i!i f:Si 

0.3 2.01 1.99 1.10 1.02 1.97 1.19 

0.4 1.3 1 1.43 1.76 1.54 0.97 0.89 1.64 1.03 

0.5 1.16 1.18 1.53 1.26 0.86 0.78 1.39 0.91 

Indicates the factors of safety corresponding to the range of estimated horizontal peak ground-surface accelerations near the 

site. 

unlikely mechanism and geometry of failure, and conservative construction. 

Because seismic failures of soil-nailed excavations have 

not occurred in the past and are therefore absent from the 

literature, dynamic centrifuge testing was performed to provide evidence of the most probable failure mechanism. The 

centrifuge testing revealed that the most likely failure mechanism is the German type of failure mechanism. Furthermore, 

a simple analysis of the dynamic centrifuge test results and 

field observations showed that the Caltrans method for calculating the factor of safety, which also incorporates the 

German type of failure mechanism, yields very consistent 

and logical results. Accordingly, the Caltrans method implemented by the computer program SNAIL seems to be an 

appropriate method for calculating the static and dynamic 

stability of grouted soil-nailed excavations of the type discussed in this paper. 

ACKNOWLEDGMENTS 

The investigations described in this paper were supported 

by the National Science Foundation through the grants BCS9001610, BCS-9011819 and BCS-9224479. This support is 

gratefully acknowledged. We would also like to thank professors R. Dobry, A.-W. Elgamal, and T.F. Zimmie, Dr. P. 

Van Laak, and Dr. L. Liu, all from the RPI Geotechnical 

Centrifuge Research Center for their excellent service and 

continued assistance. Special thanks go to Dr Macan 

Doroudian, former student at UCLA, for his help in performing the centrifuge tests. We would also like to recognize Mr 

Ken Jackura from Caltrans who supplied the computer program SNAIL and provided valuable suggestions. 

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Barar, P., 1990, The behavior of five soil nailed earth retaining structures during the Loma Prieta earthquake of October 17, 1989: 

Report prepared for the Department of Civil Engineering, 

University of California, LosAngeles, 101 p. 

Bruce, D.A., and Jewell, R.A., 1986, Soil nailing-Application and 

practice, part 1 : Ground Engineering, November p. 10- 15. 

1987, Soil nailing-Application and practice, part 2: Ground 

Engineering, January, p. 21-33. 

Caltrans, 1993, California Department offransportation-SNAIL 

Computer Program, Version 2.06: California Department of 

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Chapman, K.R., and Ludwig, C., 1993,ADSC participates in FHWA 

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Elgamal, A.-W., Dobry, R., and Van Laak, P., 1991, Design construction and operation of 100 g-ton centrifuge at RPI, in Proceedings of the International Conference Centrifuge 199 1, 

Boulder: Balkema, Rotterdam, p. 27-34. 

Felio, G.Y, Vucetic, M., Hudson, M. Barai; P., and Chapman, R., 

1990, Performance of soil nailed walls during the October 17, 

1989 Loma Prieta Earthquake, in Proceedings of the FortyThird Canadian Geotechnical Conference, Quebec: Canadian 

Geotechnical Society, p. 165- 173. 

Federal Highway Administration, 1993, FHWA Tour for 

Geotechnology-Soil Nailing: Federal HighwayAdministration Report No. FHWA-PL-93-020, 84 p. 

Gassler, G., 1992, Full scale test on a nailed wall in consolidated 

clay, in Proceedings of the International Symposium on Earth 

Reinforcement Practice, Fukuoka, Kyushu, Japan: Balkema, 

Rotterdam, v. 1, p. 475-480. 

Gassler, G., and Gudehus, G., 1981, Soil nailing-some aspects of 

a new technique, in Proceedings of the 10th International Conference Soil Mechanics and Foundation Engineering: Balkema, 

Rotterdam, v. 3, p. 665-670. 

Guilloux, A., and Schlosser, F., 1982, Soil nailing practical applications, in Proceedings of the Symposium on SoilIRock Improvement Technique: A.I.T. 

Hudson, M.B., 1990,The effects of the Loma Prieta earthquake on 

soil-nailed earth retaining structures in the San Francisco Bay 

area: Civil Engineering Department, University of California, 

Los Angeles, M.S. thesis, 108 p. 

Japan Highway Public Corporation, 1987, Guide for design and 

construction on reinforced slope with steel bars: September 

33 p. 
ANALYSIS OF SOIL-NAILED EXCAVATIONS STABILITY DURING THE 1989 LOMA PRIETA EARTHQUAKE D45 

Jewell, R.A., and Pedley, M.J., 1992,Analysis for soil reinforcement with bending stiffness: Journal of Geotechnical Engineering, v. 118, no. 10, p. 1505-1528. 

Juran, I., Baudrand, G., Farrag, K., and Elias,V., 1990, Kinematical limit analysis for soil nailed structures: Journal of 

Geotechnical Engineering, v 116, no. 1, p. 54-72. 

Juran, I., and Elias, V., 199 1, Ground anchors and soil nails retaining structures, in Foundation Engineering Handbook (2d ed.): 

Van Nostrand Reinhold, New York, p. 868-905. 

Koerner, R.M., 1984, Construction and geotechnical methods in 

foundation engineering: McGraw-Hill, Inc. 

Lambe, PC., and Jayaratne, N.N., 1987, Construction of retaining 

walls from the top down: Center forTransportation Engineering Studies, Department of Civil Engineering, North Carolina 

State University at Raleigh, Research Report No. 2324 1-86- 

6, 183 p. 

Mitchell, J.K., and Villet, W.C.B., 1987, Reinforcement of Earth 

Slopes and Embankments,Appendix C, Soil Nailing: National 

Cooperative Highway Research Program, Report 290,Washington, D.C., June, p. 258-323. 

Myles, B., and Bridle, R.J., 1991, Fired soil nails-the machine: 

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Ochiai, H., Hayashi, S., and Otani, J., eds, 1992, Proceedings of 

the International Symposium on Earth Reinforcement Practice, IS Kyusha '92, v. 1 and 2. Fukuoka, Japan: Balkema, 

Rotterdam. 

Plumelle, C., Schlossei; F., Delage, l?, and Knochenmus, G., 1990, 

French national research project on soil nailing Clouterre: 

American Society of Civil Engineers SpecialTechnical Publication No. 25, p. 660-675. 

Schlosser, F., Jacobsen, H.M., and Juran, I., 1983, Soil reinforcement-general report, in Proceedings of the Eighth European 

Conference on Soil Mechanics and Foundation Engineering: 

Balkema, Rotterdam, v. 3, p. 1159-1180. 

Seed, R.B., Dickenson, S.E., and Idriss, I.M., 1991, Principal 

geotechnical aspects of the 1989 Loma Prieta earthquake: 

Soils and Foundation, v. 3 1, no. 1, p. 1-26. 

Shen, C.K., Herrmann, L.R., Romstad, K.M., Bang, S., Kim,Y.S., 

and DeNatale, J.S., 1981,An in situ earth reinforcement lateral support system: Department of Civil Engineering, University of California, Davis, Report No. 81-03, 188 p. 

Stocker, M.F., Korber, G.W., Gassier, G., and Gudehus, G., 1979, 

Soil nailing, in Proceedings of the International Conference 

on Soil Reinforcement, Paris, v 2, p. 469-474. 

Stocker, M.F., and Riedingeg G., 1990, The bearing behavior of 

nailed retaining structures: American Society of Civil Engineers Special Technical Publication No. 25, p. 61 2-628. 

Tatsuoka, F., 1992, Keynote lecture-roles of facing rigidity in soil 

reinforcing, in Preprints of Special and Keynote Lectures of 

the International Symposium on Earth Reinforcement Practice Kyusha '92, Fukuoka, Japan: Institute of Industrial Science, Tokyo, p. 77-115. 

Tufenkjian, M.R., andvucetic, M., 1992, Seismic stability of soil 

nailed excavations, in Proceedings of the International Symposium on Earth Reinforcement Practice, Fukuoka, Kyushu, 

Japan: Balkema, Rotterdam, p. 573-578. 

1993, Seismic stability of soil nailed excavations: Civil and 

Environmental Engineering Department, University of California, Los Angeles, Research Report ENG-97- 169, June, 298 

P 

Tufenkjian, M.R.,Vucetic, M., and Doroudian, M., 1991, Stability 

of soil nailed excavations, in Proceedings of the International 

Workshop on Technology for Hong Kong's Infrastructure Development: Commercial Press Ltd., Hong Kong, p. 751-762. 

Vucetic, M., Iskanda~ V.E., Doroudian, M., and Luccioni, L., 1996, 

Dynamic failure of soil-nailed excavations in centrifuge,in 

Proceedings of the International Symposium on Earth Reinforcement Practice, Fukuoka, Kyushu, Japan: Balkema, 

Rotterdam, v. 1, p. 829-834. 

Vucetic, M.,Tufenkjian, M.R., and Doroudian, M., 1993, Dynamic 

centrifuge testing of soil nailed excavations: Geotechnical 

Testing Journal, v. 16, no. 2, p. 172-187. 
THE LOMA PRIETA, CALIFORNIA, EARTHQUAKE OF OCTOBER 17,1989: 

PERFORMANCE OF THE BUILT ENVIRONMENT 

EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

EMPIRICAL ANALYSIS OF PEAK HORIZONTAL ACCELERATIO 'N, 

PEAK HORIZONTAL VELOCITY, AND MODIFIED MERCALLI INTENSITY 

CONTENTS 

By Kenneth W. Campbell, 

EQE International, Inc., Evergreen, Colo. 

Page 

D47 

47 

48 

48 

48 

55 

55 

56 

57 

57 

59 

61 

61 

63 

64 

66 

66 

ABSTRACT 

A total of 137 ground-level accelerograms recorded during the earthquake were used to develop empirical models for predicting the dependence of peak horizontal 

acceleration (PHA), peak horizontal velocity (PHV), and 

modified Mercalli intensity (MMI) on distance from the 

fault, source-to-site azimuth, and surficial geology. These 

models indicate that (1) PHA recorded on alluvium at 

distances farther than 50 km from the seismogenic rupture zone were significantly higher than those predicted 

from attenuation relationships available at the time of 

the earthquake; (2) both the amplitude and rate of attenuation of PHA, PHV, and MMI exhibited a strong 

dependence on azimuth with ground motions recorded 

at azimuths corresponding to Santa Cruz, San Francisco, 

and Oakland (west and northwest direction) having significantly higher amplitudes and lower rates of attenuation than those recorded in the east and northeast direction; and (3) PHA recorded at intermediate and far distances at all azimuths exhibited a strong dependence on 

site geology, with PHA on bay mud (SC-IV, Sa), alluvium 

(SC-111, SD), and soft rock (SC-11, Sc) being on average 

2.76, 1.47, and 1.25 times higher than on hard, or predominantly Franciscan, rock (SC-Ib, SB). The azimuthal 

dependence of PHA and PHV is consistent with the observed geographic distribution of MMI. The observed 

azimuthal dependence of PHA, PHV, and MMI is consistent with the combined effects of source directivity, 

shear-wave radiation pattern, and propagation effects (that 

is, critical reflections from layers within and at the base 

of the crust). The observed dependence of PHA on 

surficial geology is consistent with low-strain site amplification factors recommended for use in building-code 

and engineering applications. However, these recommended factors are generally not consistent with the 

amplitude-dependent factors predicted from empirical 

attenuation relationships. A comparison of the model predictions with published data indicates that the models 

accurately predict the observed dependence of PHA, PHI( 

and MMI on azimuth and distance throughout central 

California to distances as far as 250 km from the earthquake rupture zone. 

INTRODUCTION 

The earthquake caused considerable damage to man-made 

structures in the San Francisco Bay area. The amount of 

damage observed in San Francisco and Oakland during this 

earthquake is superseded only by that sustained during the 

great 1906 San Francisco earthquake (Housner, 1990; U.S. 

Geological Survey Staff, 1990). The amount of damage was 

unusually large compared to other California earthquakes 

of similar magnitude, especially when one considers the 

relatively short duration of the ground motions and the 

relatively large epicentral distance to San Francisco and 

Oakland. This damage is, however, consistent with the 

unusually high ground motions that were recorded in San 

Francisco and Oakland. 

A compilation of peak accelerations published by Boore 

and others (1989) and Brady and Shakal(1994) indicates that 

peak horizontal accelerations as large as 0.20 to 0.25 g were 

routinely recorded on alluvium and soft soil in the San 
D48 PERFORMANCE OF THE BUILT ENVIRONMENT 

Francisco and Oakland areas-the largest being 0.41 g on 

Holocene bay deposits (Bay mud) in a hanger at the Alameda 

Naval Air Station approximately 93 krn from the epicenter. 

Campbell (1991) demonstrated that these accelerations are 

1.5 to 3 times larger, on average, than those predicted from 

strong-motion attenuation relationships that were available 

at the time of the earthquake, but they are consistent with 

the unusually large amount of damage (Housner, 1990) and 

unusually high intensities (Stover and others, 1990) observed 

in these areas. 

In an attempt to understand this phenomenon, values 

of peak horizontal acceleration (PHA), peak horizontal 

velocity (PHV), and Modified Mercalli intensity (MMI) 

recorded and observed during this earthquake were 

analyzed to determine their dependence on distance, 

azimuth, and surficial geology. The dependence on 

distance, or what is simply referred to as "attenuation7' 

throughout this paper, includes the effects of geometrical 

spreading, material damping, and scattering. The analysis 

was intended to be an engineering rather than a 

seismological study, emphasizing those aspects of ground 

motion which have the greatest influence on the empirical 

prediction of ground motion for seismic design. As 

demonstrated later, these results clearly show why damage 

in San Francisco and Oakland was more severe than at 

other locations of comparable distance throughout the 

region, an observation consistent with the geographic 

distribution of MMI in the Bay area (Stover and others, 

1990). 

The seismological aspects of the earthquake have been 

discussed extensively in several special volumes dedicated 

to this earthquake, including Geophysical Research Letters, 

v. 17, no. 9, August, 1990; the Bulletin of the Seismological 

Society of America, v. 81, no. 5, October 1991; and companion 

volumes of this Professional Paper 1550 on "Earthquake 

Occurrence" and Professional Paper 1551 on "Strong Ground 

Motion and Ground Failure". This information will not be 

presented in this paper; rather, the reader is referred to these 

publications for a complete discussion of the seismological 

aspects of the earthquake 

STRONG-MOTION DATA 

The strong-motion database used in this study consisted 

of 137 ground-level accelerograms compiled by Campbell 

(1991). These recordings were taken from reports published 

by Maley and others (1989), Shakal and others (1989), 

Brady and Mork (1990), and California Strong Motion 

Instrumentation Program (1991a, b, c). MMI observations 

were taken from compilations by Stover and others (1990) 

and C. Stover (written commun., 1991). 

The source-to-site distance measure used in this study 

is the closest distance to the zone of seismogenic rupture, 

a measure first proposed by Campbell (1987) and later 

adopted by Campbell and Bozorgnia (1994b) and 

Campbell (1997). It is defined as that portion of the 

rupture zone responsible for most, if not all, of the 

seismically radiated energy observed at the ground 

surface. This zone was interpreted to be approximately 

40 km long, extend between depths of 4.5 to 19 km, and 

dip approximately 70' to the southwest as defined by the 

location of aftershocks that occurred within several weeks 

after the earthquake (see cross sections in Plafker and 

Galloway, 1989) and by seismological, geological, and 

geodetic studies conducted within several years after the 

earthquake (for example, Bulletin of the Seismological 

Society of America, v. 81, no. 5; Professional Paper 1551). 

A description of surficial geology for each of the recording sites was obtained from Campbell (1989, 1990), Shakal 

and others (1989), and Fumal(1991). These descriptions were 

independently verified and supplemented with data derived 

from 30 geology maps of the region. Based on these descriptions, each site was assigned to one of four surficial 

geologic classifications defined by Campbell (1981): soft soil 

(artificial fill; Holocene fluvial, bay, and estuarine deposits; and other soft soils), alluvium (Quaternary deposits >10 

m deep), soft rock (primarily Tertiary and Cretaceous sedimentary rock), and hard rock (primarily crystalline, metamorphic, and pre-Cretaceous sedimentary rock). The approximate correspondence between this site-classification 

criteria and those proposed for building-code and engineering applications by Borcherdt (1994), Borcherdt and 

Glassmoyer (1994), and the International Conference of 

Building Officials (1996) is given in table 1. 

GROUND-MOTION MODELS 

Boore and others (1989), Plafker and Galloway (1989), 

Housner (1990), and Borcherdt and Glassmoyer (1992, 

1994) found that surficial geology had a profound effect 

on peak accelerations recorded during the earthquake. 

In order to minimize this effect on the determination of 

attenuation, the regression analyses presented below were 

restricted to the 71 alluvial recordings compiled by 

Campbell (1991). These data are summarized in tables 2 

and 3. Model coefficients were estimated from a nonlinear least-squares regression algorithm developed by More 

and others (1980). 

PEAK GROUND ACCELERATION 

A preliminary analysis of PHA using an attenuation relationship whose functional form predicted a monotonically 

decreasing amplitude with distance resulted in a set of residuals that were strongly biased with respect to distance. A 

similar bias was noted by several other investigators (see 

fig. 6 of Boore and others, 1989). For distances less than 50 
EMPIRICAL ANALYSIS 

Table 1 .-Summary of site classification criteria 

[VStio, average shear-wave velocity in the upper 30 m of the deposit] 

This study 

(Campbell, 198 1, 1997) 

Borcherdt (1 994b), Proposed 1997 UBC 

Borcherdt and Glassmoyer (1994) (International Conference of Building 

Officials1 

Class Description Class Description V. Class Description 'S.30 

(mfs) (rnls) . , 

Hard rock Crystalline, metamorphic, and pre- SC-Ib Firm to hard rock 700-1,400 So Rock 750-1,500 

Cretaceous sedimentary rock 

Soft rock Tertiary and Cretaceous sedimentary SC-11 Gravelly soils 375-700 Sc Very dense soil 360-760 

rock, soft volcanics and soft to firm and soft rock 

rock 

Alluvium Unconsolidated Quaternary deposits SC-111 Stiff clays and 200-375 Sn Stiff soil 180-360 > 1 Om deep sandy soils 

Soft soil Artificial fill and Holocene fluvial, SC-1V Soft soil 100-200 Sc Soft soil 480 

bay, and estuarine deposits (for 

example, Bay mud') 

~efers toHolocene bay mud deposits along the margin of San Francisco Bay. 

Table 2.-Strong-motion data on alluvium (SC-III, Sn)' used for regression analyses of PHA 

[PHA, mean peak horizontal acceleration; RÃ closest distance to seismogenic rupture zone; g, acceleration of gravity] 

Station No. 

1103 

1116 

1226 

1227 

1230 

1239 

1265 

1439 

1446 

1474 

1479 

1481 

575 

1601 

1652A 

1652C 

1656 

1675 

1678 

1686 

1687 

1689 

1695 

46 173 

47125 

47179 

47288 

47380 

47381 

47459 

47460 

47524 

56012 

57064 

57066 

57191 

PHA 

(g) 

0.098 

0.125 

0.005 

0.360 

0.195 

0.1 10 

0.050 

0.095 

0.095 

0.070 

0.1 15 

0.165 

0.240 

0.240 

0.255 

0.205 

0.280 

0.090 

0.180 

0.175 

0.105 

0.085 

0.205 

0.065 

0.505 

0.105 

0.050 

0.350 

0.460 

0.333 

0.080 

0.280 

0.050 

0.120 

0.165 

0.120 

-- 

Station No. RS 

(km) 

57355 18.6 

57356 18.9 

57357 19.6 

57382 15.0 

57425 23.2 

57458 81.0 

57476 11.4 

57502 27.6 

57504 20.7 

57528 64.8 

57562 13.4 

57563 13.2 

58000 77.8 

58065 8.2 

58224 71.5 

58233 48.9 

58235 8.4 

58261 64.4 

58393 51.6 

58394 60.7 

58462 49.5 

58483 70.4 

58490 60.8 

58492 86.6 

58496 76.3 

58498 53.1 

58501 53.2 

58503 87.4 

58505 87.1 

68003 117.0 

68150 121.1 

68387 150.8 

68489 150.5 

6849 1 150.5 

69039 154.0 

~efinition of site-classification criteria is given in table 1. 
PERFORMANCE OF THE BUILT ENVIRONMENT 

Table 3.-Strong-motion stations on alluvium (SC-111, SJ used in regression 

analyses of PHA 

[PHA, mean peak horizontal acceleration. Owner: USGS, U.S. Geological Survey; CDMG, California 

Division of Mines and Geology. Location: BSMT, basement; GRND, ground level; TOE, toe of dam] 

-- - -- 

Station No. Owner Station name Location 

1 103 USGS BERKELEY-2168 SHATTUCK, EAST BSMT 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

USGS 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

CDMG 

SAN FRANCISCO-CSUSF, THORTON HALL 

LIVERMORE-VA HOSPITAL, BLDG. 62 

PAL0 ALTO-VA HOSPITAL, BLDG. 1 

MENLO PARK-VA HOSPITAL, BLDG. 37 

SAN FRANCISCO-TRANSAMERICA TOWER 

DEL VALLE DAM 

RICHMOND-BULK MAIL CENTER 

SAN FRANCISCO-STANDARD OIL BUILDING 

BEAR VALLEY #5-CALLENS RANCH 

BEAR VALLEY # 1 &WEBB RESIDENCE 

BEAR VALLEY # 12-WILLIAMS RANCH 

HOLLISTERÃ‘CIT HALL ANNEX 

STANFORD UNIVERSITY-SLAC TEST LAB 

ANDERSON DAM-DOWNSTREAM 

ANDERSON DAM-TOE 

HOLLISTER-DIFFERENTIAL ARRAY (SMA) 

SAN FRANCISCO-1 295 SHAFTER 

SAN FRANCISCOÃ‘GOLDE GATE BRIDGE 

FREMONT-EMERSON COURT 

CALAVERAS RESERVOIR-SOUTH 

DUBLIN-FIRE STATION 

SUNNYVALE-COLTON AVENUE 

BITTERWATERÃ‘COALING ROAD 

CAPITOLA-FIRE STATION 

SALINAS-JOHN & WORK 

SAN BENITO 

GILROY #2-HIGHWAY 101 MOTEL 

GILROY #3-SEWAGE PLANT 

WATSONVILLE-TELEPHONE BUILDING, N.W. 

GREENFIELD-POLICE STATION 

HOLLISTER-GLORIETTA WAREHOUSE, F.F. 

LOS BANOS 

FREMONT-MISSION SAN JOSE 

AGNEWS-AGNEWS STATE HOSPITAL 

HALLS VALLEY-GRANT PARK 

SAN JOSEÃ‘GREA WESTERN BUILDING, SOUTH 

SAN JOSE-TOWN PARK TOWERS, SOUTH 

SAN JOSE-SANTA CLARA COUNTY BLDG., S.W. 

GILROY #4-SAN YSIDRO SCHOOL 

GILROY #7-MANTELLI RANCH 

TRACY-SEWAGE PLANT 

GILROYÃ‘OL FIREHOUSE 

MILPITAS-2-STORY BUILDING, EAST 

COYOTE LAKE DAM-DOWNSTREAM 

LIVERMORE-FAGUNDES RANCH 

SAN JOSE-3-STORY BUILDING, EAST 

SAN JOSE-SANTA TERESA HILLS 

BERKELEY-UCB, STADIUM GROUNDS 

SARATOGA-1 4675 ALOHA 

OAKLANDÃ‘TITL & TRUST BUILDING, N.E. 

LOWER CRYSTAL SPRINGS DAM-DOWNSTREAM 

SARATOGA-SCHOOL GYM, SHEAR WALL, NO. 

SO SAN FRANCISCO-KAISER MEDICAL CENTER 

HAYWARD-MUIR SCHOOL (APEEL #2E) 

SAN BRUNO-U.S. POSTAL BUILDING, CENTER 

HAYWARDÃ‘6-STOR OFFICE BUILDING, CENTER 

OAKLANDÃ‘24-STOR BUILDING, WEST 

SAN BRUNOÃ‘6-STOR OFFICE BUILDING 

CONCORDÃ‘8-STOR BUILDING, CENTER 

BERKELEY-2-STORY HOSPITAL 

HAYWARD-BART STATION, PARKING LOT 

HAYWARD-BART ELEVATED SECTION, BENT 132 

RICHMONDÃ‘3-STOR GOVERNMENT BLDG., E. 

RICHMONDÃ‘CIT HALL, PARKING LOT 

OLEMA-POINT REYES RANGER STATION 

NAPA-NAPA COLLEGE 

SANTA ROSA-5-STORY BUILDING, WEST 

SANTA ROSA-14-STORY BUILDING, CENTER 

SANTA ROSA-HENDLEY & TUPPER 

GRND 

BSMT 

BSMT 

GRND 

BSMT 

TOE 

GRND 

BSMT 

GRND 

GRND 

GRND 

BSMT 

GRND 

GRND 

TOE 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

BSMT 

GRND 

BSMT 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

BSMT 

GRND 

GRND 

BSMT 

GRND 

GRND 

GRND 

BSMT 

GRND 

GRND 

BSMT 

GRND 

GRND 

GRND 

GRND 

GRND 

GRND 

69039 CDMG BODEGA HEAD-DORAN BEACH GRND 

~efinition of site-classification criteria is given in table 1. 
EMPIRICAL ANALYSIS D5 1 

km from the seismogenic rupture zone, this preliminary 

model tended to overestimate PHA, whereas for distances 

greater than 80 km the opposite effect was observed. Between these two distances the residuals showed a gradual 

transition from one trend to the other. Upon further analysis it became clear that this bias was caused by the presence of a zone of nearly constant acceleration between 

distances of about 50 and 80 km from the seismogenic 

rupture zone. 

In order to model the observed behavior, a second regression analysis was performed in which the distance limits 

defining the zone of constant acceleration were included as 

regression coefficients. The only constraint imposed in this 

analysis was that PHA remain constant between these two 

distance limits. This resulted in the following tripartite attenuation relationship: 

In PHA = 1.876 -1 .03 In [Rs + 7.791, Rs 2 50.6 km, 

= -2.313, 50.6<Rs<79.1 km, (1) 

= 3.223-1.24 In [Rs + 7.791, Rs2 79.1 krn; 

where PHA is the arithmetic mean of the two horizontal 

components of peak ground acceleration in fractions of gravity (g = 981 cmls), and Rs is the closest distance to the zone 

of seismogenic rupture in kilometers. The standard error of 

estimate, qPHA, associated with this model was 0.406. This 

relationship together with its supporting data is plotted in 

figure 1. This figure, which is plotted in terms of a linear 

rather than a logarithmic distance scale to accentuate the 

larger distances, clearly shows the existence of a zone of 

nearly constant acceleration between distances of about 50 

and 80 km. 

A comparison of equation (1) with empirical attenuation 

relationships available in 1990 was made by Campbell 

(1991). Because of what was then considered unusual attenuation characteristics exhibited by the earthquake, most 

of these attenuation relationships were found to underpredict 

systematically the accelerations by as much as a factor of 

1.5 to 3 at distances greater than about 50 krn. However, 

they all accurately predicted the accelerations at distances 

of 50 km and less. Similar behavior had been previously observed during the 1987 Whittier Narrows, California, earthquake (Campbell, 1988) and was subsequently observed 

during the 1992 Landers, California, earthquake (Campbell 

and B ozorgnia, 1994a) 

An analysis of residuals conducted by Campbell (1991) 

using equation (1) indicated that there was a strong dependence of the residuals on azimuth. This dependence is summarized in table 4. Sites located northwest of the epicenter 

at azimuths of 320' to 350' (measured clockwise from north 

with respect to the epicenter) were found to have median 

accelerations that were 19 percent higher, and sites located north and northeast of the epicenter at azimuths of 

350' to 030' had median accelerations that were 29 percent lower than those given by equation (1). The rate of 

attenuation of PHA was found to be lower toward the 

northwest and higher toward the northeast as compared to 

the average rate of attenuation given by equation (1). These 

differences were found to be significant at the 90 percent 

confidence level. 

Based on the above results, a third regression analysis was 

performed in which the rate of attenuation was allowed to 

vary as a function of azimuth. Since there was an insufficient number of strong-motion recordings with which to 

0 20 40 60 80 100 120 140 160 

Distance to Seismogenic Rupture (km) 

Figure 1 .-Attenuation relationship for the earthquake developed by Campbell (1991): solid line, 

median; dashed lines, one standard error bounds; solid squares, recorded values of mean peak 

horizontal acceleration. 
D52 PERFORMANCE OF THE BUILT ENVIRONMENT 

Table 4.-Summary of relative azimuthal effects for PHA 

[PHA, mean peak horizontal acceleration; -, no value] 

Azimuth Direction from Relative factor 

(O clockwise from N.) epicenter Median 68% confidence 

interval 

1989 Lorna Prieta earthquake (this study) 

350-030 Northeast 1 .OO 0.73-1.36 ' 

320-350 Northwest 1.68 1.19-2.38 -- - -- 

1992 Landers earthquake (Campbell and Bozorgnia, 1994a) 

175-240 Southwest 1 .OO - 

240-300 West 1.29 - 

300-0 1 5 Northwest 1.62 - 

115-175 Southeast 1.59 

constrain the rate of attenuation at some azimuths and 

distances, the abundant observations of MMI on alluvium 

were used to supplement the limited strong-motion data 

in these regions. The procedure for doing this is discussed 

below. The resulting attenuation relationship is given by the 

expression, 

In PHA =In PHAi, 

= W In PHAl+ (1-W) In PHA2, 

= In PHA2, 

= Win PHA2 + (1-W) In PHA3, 

= In PHA3, 

= Win PHA3 + (1-W) In PHA4, 

= In PHA4, 

= W In PHA4 + (1-W) In PHAi, 

where 

(2) 

w = '/2 (1 + cos S), 

PHA and Rs are as defined in equation (I), 6, is sourceto-site azimuth measured clockwise from north (assuming an epicentral location at 37,037' N. and 

121.883' W.), 0. is the beginning azimuth for the 

specified range, â‚¬ is the ending azimuth for the specified range, and W is an azimuthally dependent weighting factor. The standard error of estimate of this model 

is clnpHA = 0.297, which corresponds to a 27 percent 

reduction in the standard error associated with equation (1). This relationship, along with its supporting 

data, is plotted in figures 2 through 6. 

Equation (2) was developed from a nonlinear regression analysis in which the amplitude of PHA at Re= 0, 

the assumed source of the radiated energy, was constrained to be constant at all azimuths. The range of azimuths used to define the various segments of the model 

were determined from the azimuthal dependence of PHA 

observed by Campbell (199 1) and from maps of peak acceleration and MMI published by Maley and others 

(1989), Shakal and others (1989), and Stover and others 

(1990). 

Intensity observations on alluvium compiled by Stover 

and others (1990) and C. Stover (written commun., 1991) 

were used to constrain the attenuation of PHA at distances 

beyond about 75 km, where the number of strong-motion recordings was limited. This was done through an 

iterative procedure in which azimuthally dependent estimates of R, and PHA for MMI contours IV, V, VI, VII, 

and VIII were used to simultaneously develop an attenuation relationship for PHA and an earthquake-specific 

relationship between PHA and MMI (see next section). 

Initial estimates of PHA for MMI contours IV, V, and VI 

were taken from a relationship developed by Trifunac 

(1976), and those for MMI contours VII and VIII were 

obtained from a preliminary version of equation (2). 

The steps involved in this iterative procedure are summarized as follows: (1) estimates of PHA and Re for MMI contours IV, V, and VI were used in conjunction with the strongmotion recordings compiled for this study to develop an attenuation relationship for PHA, (2) estimates of PHA for 

distances associated with MMI contours IV through VIII 

obtained from the attenuation relationship developed in step 

1 were used to develop a relationship between PHA and MMI, 

and (3) steps 1 and 2 were repeated until an acceptable level 

of convergence between the PHA attenuation relationship 

and the PHA-MMI relationship was obtained. 
EMPIRICAL ANALYSIS D53 

1 10 100 1000 

Distance to Seismogenic Rupture (km) 

Figure 2.-Attenuation relationship developed in this study for alluvial sites located at epicenter-to-site azimuths of 270' to 340' (northwest direction) measured clockwise from north. 

F l v Corralitos 

t 1 ' Recorded I 

1 10 100 1000 

Distance to Seismogenic Rupture (km) 

Figure 3.-Attenuation relationship developed in this study for alluvial sites located at epicenter-to-site azimuths of 000' to 060' (northeast direction) measured clockwise from north. 
PERFORMANCE OF THE BUILT ENVIRONMENT 

1 10 100 10 00 

Distance to Seismogenic Rupture (km) 

Figure 4.-Attenuation relationship developed in this study for alluvial sites located at epicenter-to-site azimuths of 095' to 195' (southeast direction) measured clockwise from north. 

[ Alluvium: 

10 

Distance to Seismogenic Rupture (km) 

\ 

Figure 5.-Attenuation relationship developed in this study for alluvial sites located at epicenter-to-site azimuths of 210' to 255' (southwest direction) measured clockwise from north. 

'V Corralitos 

a Recorded 
EMPIRICAL ANALYSIS 

h 

- 

- 

- .------ - ------------ - 

- 

- 

- - 

- 

- 

- 

- 

- 

- Azimuth: 

- North west (270-340) - 

- 

- - Northeast (000-060) 

. . . . . . . . . . . . Southeast (095-196) 

- Southwest (210-266) 

10 100 10 00 

Distance to Seismogenic Rupture (km) 

Figure 6.-Comparison of attenuation relationships developed in this study for alluvial sites for epicenter-to-site azimuths corresponding 

to the northwest, northeast, southeast, and southwest directions. 

MODIFIED MERCALLI INTENSITY 

The relationship between MMI and PHA obtained using 

the iterative procedure described in the previous section is 

given by the expression, 

MMI = 8.76 + 0.637 (In PHA) - 0.114 (In PHA)2, 

which is considered to be valid for accelerations ranging from 

0.014 to 1.0 g. This relationship is plotted in figure 7, where 

it is compared to the relationship proposed by Trifunac (1976) 

for California earthquakes. The Trifunac relationship has been 

adjusted by +% MMI to predict the value of PHA at the 

MMI contour rather than at the middle of the region between 

contours. Figure 7 indicates that estimates of MMI from equation (3) are consistently lower than those predicted by the 

Trifunac relationship. EQE International, Inc. (1995) found a 

similar discrepancy between observed and predicted intensities for the 1994 Northridge earthquake. 

The standard error of estimate, (oMMI, associated with equation (3) is 0.107, indicating a remarkably strong correlation 

between PHA and the MMI contour value. However, for 

purposes of prediction, it is the error associated with the estimation of MMI at an individual observation site, not the 

error associated with a specific MMI contour that is of interest. An estimate of this latter error was obtained from the 

standard deviation of the computed differences between the 

integer representations of the observed and predicted intensities at each alluvial site for which there was an MMI observation (C. Stover, written commun., 1991). This analysis 

indicated that the standard error associated with the estimation of MMI at an individual site is about 0.55. 

PEAK HORIZONTAL VELOCITY 

Because there was an insufficient number of digitized recordings, it was not possible to develop an independent attenuation relationship for peak horizontal velocity for alluvial sites. 

Instead, the available processed recordings were used to develop the following relationship between PHV and PHA:, 

In PHV = ln PHVi , 095' < 0 < 340' 

= W In PHVl + (1-W) In PHV2, 340' < 0 < 360' 

= In PHV2, 000' 5 0 < 060'; 

= W In PHV2 + (1-W) In PHVl, 060' < 0 < 095'; 

where (4) 

In PHV; = 4.720 + In PHA, i 1; 

= 4.551 + ln PHA, i 2, 

PHV is the arithmetic mean of the two horizontal components of peak ground velocity in centimeters per second, PHA 

is calculated by equation (2), and the standard error of estimate, G,~ puv, is 0.305. The ratio of PHV to PHA given by 

equation (4) is between 94.7 and 112.1 cm/s/g, the lower 

estimate corresponding to sites located northeast of the epicenter. The higher ratios at sites located toward the southeast, west, and northwest are consistent with the effects of 
D56 PERFORMANCE OF THE BUILT ENVIRONMENT 

10, / 

Alluvium: 

- Lorna Prieta 

.--- Trifunac (1976) 

0.1 1 

Peak Horizontal Acceleration (a) 

Figure 7.Ã‘Compariso of the relationship between the Modified Mercalli intensity (MMI) contour and mean peak horizontal acceleration 

(PHA) developed in this study for alluvial sites with a similar relationship developed by Trifunac (1976) for California earthquakes. The Trifunac 

relationship has been adjusted by +I12 MMI to predict the value of PHA at the MMI contour rather than at the middle of the region between contours. 

directivity and shear-wave radiation pattern that have a larger 

impact on the longer-period ground motions associated with 

PHV than on the shorter-period motions associated with PHA. 

The ratios given by Equation (4) are summarized in Table 5, 

where they are compared to PHVIPHA ratios recommended 

by Mohraz and Elghadamsi (1989). 

EFFECT OF SURFICIAL GEOLOGY 

An analysis of residuals was used to identify the potential 

effects of surficial geology on the observed values of PHA 

(Campbell, 1991). The residuals, which are referred to by 

the notation NR in the discussion that follows, were computed with respect to the predicted value of In PHA from 

equation (1) and divided by CT~~~~~. A positive residual indicates that the recorded value is larger than that predicted by 

the model, and a negative residual indicates that the recorded 

value is smaller than that predicted by the model. Since the 

residuals have been normalized by G~~~~~, an NR of kl represents a deviation of plus-and-minus one standard error between the recorded and predicted value of In PHA. If the 

residuals are assumed to be normally distributed, NR beTable 5.-Summary of PHV to PHA ratios for alluvium (SC-111, SD)' 

[PHV, mean peak horizontal velocity; PHA, mean peak horizontal acceleration; g, acceleration of gravity; -, no value] 

Azimuth 

(O clockwise from N.) 

Soil depth PHV/PHA ratio (cm/s/g) 

(feet) Median 68% confidence 

interval - 

1989 Loma Prieta earthquake (this study) 

000-060 - 95 70-1 29 

095-340 - 112 83-1 52 

Mohraz and Elghadamsi (1989) 

- <30 88 51-151 

- 30-200 84 53-133 

- >200 133 90-1 96 

1 Definition of site-classification criteria is given in table 1. 
EMPIRICAL ANALYSIS D57 

comes an estimate of the standard normal variate. In this 

case a normalized residual of 0 corresponds to a recording 

that falls at the 50th percentile (the median) of the distribution of residuals, and normalized residuals of - 1 and +1 correspond to recordings that fall at the 16th percentile and 84th 

percentile of this distribution, respectively. The correspondence between the site classification criteria given below and 

those recommended for building-code and engineering applications by Borcherdt (1994), Borcherdt and Glassmoyer 

(1994), and the International Conference of Building Officials (1996) is given in table 1. 

ALLUVIUM 

A plot of NR versus Rsfor alluvium (SC-111, Sn) is displayed in figure 8. There is no visible trend in this plot. This 

is expected since these recordings were used to develop the 

relationship. The diamonds in this figure represent recordings from embedded sites, primarily from basements of buildings. A hypothesis test indicated that the embedded recordings had a mean residual close to zero, no different from the 

mean residual of the ground-level and free-field recordings 

at the 90 percent confidence level. This test appears to contradict the results of previous studies which have found embedded sites to have smaller accelerations than nonembedded 

sites (see Campbell, 1987, 1988, 1989, 1990); however, the 

relatively large distances associated with many of these recordings has apparently minimized this effect (Campbell, 

1987). Figure 9 shows the distribution of residuals with respect to Rs and azimuth. This plot indicates that distances 

greater than 80 km between azimuths of 000' and 160' and 

all distances between azimuths of 160' and 320' are poorly 

represented by the recordings. 

A plot of NR versus azimuth is displayed in figure 10. This 

plot shows that there is a strong azimuthal bias in the residuals, with NR between azimuths of 320' and 350' biased toward positive values, representing an underestimation of PHA 

by the regression model, and those between azimuths of 350' 

and 030' biased toward negative values, representing an overestimation of PHA. The underestimation occurs at azimuths 

corresponding to San Francisco and Oakland, the overestimation at azimuths corresponding to Hayward and San Jose. 

This bias was accounted for in equation (2). Evidence presented later strongly suggests that this azimuthal bias is due 

to the combined effects of differences in attenuation, source 

directivity, and shear-wave radiation pattern. Stover and others (1990) found a similar azimuthal bias in reported values 

of MMI. 

SOFT AND HARD ROCK 

A plot of NR versus RSfor soft rock (SC-11, Sc) and hard 

rock (SC-Ib, Sv) is displayed in figure 11. This plot indicates that both soft-rock and hard-rock recordings at distances of 20 km and greater tend to have smaller accelerations than those recorded on alluvial sites. The effect 

is most significant for hard-rock sites. Hypothesis tests 

indicated that the mean residuals of both the soft-rock 

and hard-rock recordings at similar azimuths were significantly different from the mean residual of the alluvial recordings at the 90 percent confidence level. A simi0 20 40 60 80 100 120 14 0 160 

Distance to Seismogenic Rupture (km) 

Figure 8.-Plot of normalized residuals versus closest distance to the seismogenic rupture zone for 

alluvial sites: squares, ground-level and free-field recordings; diamonds, embedded recordings. 
PERFORMANCE OF THE BUILT ENVIRONMENT 

0 40 80 120 160 200 240 280 320 360 

Azimuth from Epicenter (deg) 

Figure 9.-Plot of closest distance to the seismogenic rupture zone versus epicenter-to-site azimuth 

measured clockwise from north for alluvial sites: squares, ground-level and free-field recordings; 

diamonds, embedded recordings. 

0 40 80 120 160 200 240 280 320 360 

Azimuth from Epicenter (deg) 

Figure 10.-Plot of normalized residuals versus epicenter-to-site azimuth measured clockwise from 

north for alluvial sites: squares, ground-level and free-field recordings; diamonds, embedded recordings. 

lar effect was observed during the 1987 Whittier Narrows 

earthquake (Campbell, 1988). 

Figure 12 is a plot of the distribution of residuals with 

respect to /?,and azimuth. This plot shows that, although 

rock recordings are fairly evenly distributed with respect 

to azimuth, there are few recordings at distances greater 

than 50 km except at azimuths corresponding to San Francisco and Oakland. 

A plot of NR versus azimuth is displayed in figure 13. 

The number of recordings on rock is less than for alluvium, and their residuals are biased toward negative values. As a result, it is not obvious from this figure whether 

the rock recordings show a similar bias with azimuth. 

The residuals corresponding to azimuths of 320Â°t 350' 

do appear to be larger, on average, than those corresponding to azimuths of 350' to 030Â° which lends qualitative 
EMPIRICAL ANALYSIS 

0 20 40 60 80 100 120 140 16 0 

Distance to Seismogenic Rupture (km) 

Figure 11 .-Plot of normalized residuals versus closest distance to the seismogenic rupture zone for 

rock sites: asterisks, ground-level and free-field recordings on soft rock; crosses, embedded recordings on soft rock; squares, ground-level and free-field recordings on hard rock; diamonds, embedded 

recordings on hard rock, 

0 40 80 120 160 200 240 280 320 360 

Azimuth from Epicenter (deg) 

Figure 12.-Plot of closest distance to the seismogenic rupture zone versus epicenter-to-site 

azimuth measured clockwise from north for rock sites: asterisks, ground-level and free-field 

recordings on soft rock; crosses, embedded recordings on soft rock; squares, ground-level and 

free-field recordings on hard rock; diamonds, embedded recordings on hard rock. 

support to the suggestion that this bias is a propagation have substantially larger accelerations than alluvial sites. 

effect and/or a source effect, and not a site effect. As indicated in figures 15 and 16, all of these sites lie 

between azimuths of 320' and 350'. The analysis of the 

alluvial recordings indicated that these azimuths were 

BAY MUD associated with higher than average accelerations. Therefore, the large bias associated with these sites is probA plot of NR versus Rsfor Bay mud (SC-IV, Sc) is dis- ably caused by a combination of both azimuthal and site 

played in figure 14. This plot shows that Bay mud sites effects. A hypothesis test indicated that the mean residual 
PERFORMANCE OF THE BUILT ENVIRONMENT 

0 40 80 120 160 200 240 280 320 360 

Azimuth from Epicenter (deg) 

Figure 13 .-Plot of normalized residuals versus epicenter-to-site azimuth measured clockwise from 

north for rock sites: asterisks, ground-level and free-field recordings on soft rock; crosses, embedded 

recordings on soft rock; squares, ground-level and free-field recordings on hard rock; diamonds, embedded recordings on hard rock. 

0 20 40 60 80 100 120 140 160 

Distance to Seismogenic Rupture (km) 

Figure 14.-Plot of normalized residuals versus closest distance to the seismogenic rupture zone for 

Bay mud sites: squares, ground-level and free-field recordings; diamonds, embedded recordings. 

of the Bay mud recordings was significantly different 

from the mean residual of the alluvial recordings for similar azimuths at the 90 percent confidence level. Thus, the 

large amplitude bias associated with the Bay mud sites is 

statistically significant. 

The effect of surficial geology on peak acceleration is 

summarized in table 6, where it is compared to similar 

low-strain site-amplification factors observed or predicted 

by others, as discussed in a later section. The median site 

factor from the empirical attenuation relationships is the 

geometric mean of those calculated for moment magnitudes (Mw) of 6.5 and 7.5. The 68 percent confidence interval from the empirical attenuation relationships was 

derived from the complete set of calculated values for 

both Mw 6.5 and 7.5 assuming a log-normal distribution. 

According to these results, PHA recorded at intermedi- 
EMPIRICAL ANALYSIS 

0 I I I I I I I I I 

0 40 80 120 160 200 240 280 320 360 

Azimuth from Epicenter (deg) 

Figure 15.-Plot of closest distance to the seismogenic rupture zone versus epicenter-to-site azimuth 

measured clockwise from north for Bay mud sites: squares, ground-level and free-field recordings; 

diamonds, embedded recordings, 

0 40 80 120 160 200 240 280 320 360 

Azimuth from Epicenter (deg) 

Figure 16.-Plot of normalized residuals versus epicenter-to-site azimuth measured clockwise from north 

for Bay mud sites: squares, ground-level and free-field recordings; diamonds, embedded recordings. 

ate and far distances at all azimuths during the earthquake COMPARISON WITH PREVIOUS 

exhibited a strong dependence on site geology, with PHA STUDIES 

on Bay mud (SC-IV, SE), alluvium (SC-111, SD), and soft 

rock (SC-11, Sc) being on average 2.76, 1.47, and 1.25 REGIONAL ATTENUATION 

times, respectively, higher than that on hard, predominantly Franciscan rock (SC-Ib, SB). These differences The zone of nearly constant ground motions observed in 

were all found to be significant at the 90 percent confi- figure 1 between distances of about 50 to 80 km from the 

dence level. seismogenic rupture zone when recordings at all azimuths 
D62 PERFORMANCE OF THE BUILT ENVIRONMENT 

Table 6.-Summary of relative low-strain site-amplijiication factors for PHA and short-period response spectral acceleration 

[PHA, mean peak horizontal acceleration; g, acceleration of gravity; MW, moment magnitude; -, no value] 

Reference Site classification - 

Hard rock Soft rock Alluvium Soft soil (Bay 

(SC-Ib, SB) (SC-11, SC) (SC-111, SD) mud) 

(SC-IV, SE) 

1989 Lorna Prieta earthquake (MW 6.9) 

This study (median) 0.80 1 .00 1.18 2.21 

This study (68% confidence interval) 0.53-1.21 0.73-1.37 0.78-1.78 1.62-3.02 

Boore and others (1989) 1.002 1.002 1.13 2.81 

Other earthquakes 

1987 Whittier Narrows (Campbell, 1988), MW 6.1 0.84 1.00 1.47 - 1992 Landers (Campbell and Bozorgnia, 1994a), MW 0.8 1 1 .OO 1.20 - 7 7 

Empirical attenuation relationships; hard-rock PHA = 0.05 g (MW 6.5, 7.5) 

Abrahamson and Silva (1997) - 1 .OO 1.15, 1.15 - Boore and others (1997)3 0.77,0.77 1 .00 1.31,1.31 - Campbell and Bozorgnia (1994b), Campbell (1997) 0.79,0.78 1 .00 1.21, 1.34 

Idriss (1991, 1993) - 1 .00 1.17,1.17 3.26,3.82 

Sadigh and others (1997) - 1 .00 1.15,1.34 - 

MEDIAN 0.78,0.78 1.00, 1.00 1.20,1.26 3.26,3.82 

MEDIAN (all MW) 0.78 1 .OO 1.23 3.53 

68% CONFIDENCE INERVAL (all MW) - - 1.15-1.31 - 

Building-code and engineering applications4 

Borcherdt (1994b), Borcherdt and Glassmoyer (1994) 0.77 1 .OO 1.23 1.54 

Proposed 1997 UBC (International Conference of 0.83 1 .OO 1.33 1.75 

~uiidin~ Officials, 1996) 

1Definition of site-classification criteria is given in table 1. 

2No distinction was made between soft rock and hard rock. 

3Calculated for shear-wave velocities of 250,510, and 1,070 ds for alluvium, soil rock, and hard rock, respectively. 

4Corresponds to hard-rock PHA of 0.075 g (International Conference of Building Officials, 1996) and 0.1 g (Borcherdt, 1994). 

are combined is not as evident when these ground motions 

are segregated by azimuth as in figures 2 to 6, It appears 

that varying the rate of attenuation with azimuth, as was 

done in equation (2), can explain the geographic distribution of alluvial recordings better than simply varying the 

rate of attenuation with distance for all azimuths, as was 

done in equation (1). This observation is statistically confirmed by the 27 percent reduction in the standard error associated with equation (2). 

The plot comparing the attenuation of PHA with azimuth 

(fig. 6) indicates that the highest accelerations and lowest 

rates of attenuation were located southwest of the epicenter 

in the direction of Santa Cruz and Capitola. This region is 

located on the hanging wall of the causative fault and is near 

a maximum in the shear-wave radiation pattern. As a result, 

this region was subject to strong amplification due to source 

effects. The next highest accelerations and lowest rate of attenuation was located northwest of the epicenter in the direction of San Francisco and Oakland. Seismic waves propagating in this direction were subject to amplification due to 

both source directivity and radiation pattern and were observed to have an unusually low rate of attenuation as a result of (1) high Q in the lower crust (Fletcher and Boatwright, 

1991) and (2) critical reflections from velocity discontinuities 

or strong velocity gradients within the crust and from the 

base of the crust, or Moho (Catchings and Kholer, 1997). 

The lowest accelerations and highest rates of attenuation were located northeast of the epicenter. This region 

is near a minimum in the shear-wave radiation pattern 

and has propagation paths that cross several faults and 

fault-bounded basins. As a result, seismic waves 

propagating in this direction were subject to increased 

attenuation due to scattering and abrupt changes in 

crustal velocity structure. Seismic waves propagating 

toward the southeast had acceleration amplitudes intermediate between those to the northwest and northeast. This region is near a maximum in the shear-wave 

radiation pattern and was possibly subject to source 

directivity (assuming bilateral rupture). However, these 

effects were either weaker than they were toward the 

northwest, andlor propagation to the southeast was subject to greater attenuation and scattering, possibly as a 

result of different crustal properties (see Catchings and 

Kholer, 1997). 
EMPIRICAL ANALYSIS D63 

Seismological observations of test explosions and aftershocks have indicated that critical reflections from within 

the crust and from the Moho result in a relatively low rate of 

attenuation along the San Francisco Peninsula beyond distances of a few tens of kilometers that is believed to be the 

cause of the relatively high ground motions in the San Francisco and Oakland areas during the main shock (see 

Somerville and Yoshimura, 1990; McGarr and others, 199 1 ; 

Fletcher and Boatwright, 199 1 ; Borcherdt and Glassmoyer, 

1994; Somerville and others, 1994; Stevens and Day, 1994; 

Catchings and Kholer, 1997). Figures 2 to 6 generally support these observations; however, the sparse strong-motion 

data set is not sufficient to unequivocally confirm or refute 

the specific details of these observations. 

Schneider and others (1993) used a band-limited, whitenoise stochastic simulation model to assess the effects of 

source finiteness, crustal wave propagation, and site response 

on 23 strong-motion recordings located primarily along the 

San Francisco Peninsula. They compared a simple spherical 

spreading propagation model with a seismologically more 

rigorous model, which included both direct and critically 

reflected phases, and found that the two methods resulted in 

nearly the same estimates of uncertainty. They found that 

simple spherical spreading provided a good fit to the response 

spectra computed from recordings at fault distances between 

1 and 80 km, consistent with the results of this study, and 

concluded that the effects of the crustal structure could easily be accommodated by other parameters in the model. 

LOCAL SITE CONDITIONS 

As demonstrated in table 6, the relative site amplification 

factors obtained in this study are consistent with those derived for the earthquake by Boore and others (19891, Plafker 

and Galloway (19891, and Borcherdt and Glassmoyer, 1992, 

1994); for the 1987 Whittier Narrows earthquake by 

Campbell (1988); and for the 1992 Landers earthquake by 

Campbell and Bozorgnia (1994a). They are also consistent 

with site factors predicted by empirical attenuation relationships developed by Campbell and Bozorgnia (1994b) and 

Boore and others (1997), assuming a hard-rock PHA of 0.05 

g. Although not shown in table 6, the relative amplification 

factors developed in this study are also qualitatively consistent with site factors determined from recordings in the San 

Francisco Bay region of nuclear explosions in Nevada as 

well as with regional variations in intensity for the 1906 San 

Francisco earthquake (Borcherdt and Gibbs, 1976). All of 

these results are biased toward recordings obtained at relatively large distances and/or low values of PHA and, therefore, represent relatively low in-situ strains. 

Also shown in table 6 for comparison are the relative lowstrain site-amplification factors for short-period spectral accelerations recommended for building-code and engineering applications by Borcherdt (1994), Borcherdt and 

Glassmoyer (19941, and the International Conference of 

Building Officials (1996). These site factors represent hardrock (SC-Ib, SB) accelerations of 0.1 g for the Borcherdt 

study and 0.075 g for the International Conference of Building 

Officials study. The recommended site factors are very similar 

to those observed and predicted from past earthquakes, indicating a reasonable empirical basis for these recommendations, 

Empirical analyses by Idriss (1991, 1993), Campbell and 

Bozorgnia (1994b), Abrahamson and Silva (1997), Campbell 

(19971, and Sadigh and others (1997) have shown that PHA 

and short-period spectral site-amplification factors become 

significantly smaller at near-source distances, consistent with 

the general trend of the amplitude-dependent amplification 

factors recommended by Borcherdt (19941, Borcherdt and 

Glassmoyer (19941, and the International Conference of 

Building Officials (1996). Table 7 compares relative siteamplification factors between alluvium (SC-111, SD) and soft 

rock (SC-11, Sc) predicted from these studies with those recommended for building-code and engineering applications 

for hard-rock accelerations ranging from 0.05 to 0.4 g. For 

those studies which do not distinguish between soft and hard 

rock, rock has been classified as soft rock, consistent with 

the mean shear-wave velocity of 618 ds determined for 

strong-motion recording sites classified as generic rock by 

Boore and Joyner (1997). The soft-rock PHA corresponding 

to a given value of hard-rock PHA was estimated from the 

attenuation relationships of Campbell and Bozorgnia (1994b) 

and Campbell (1997). 

The results in table 7 indicate that the amplitude-dependent amplification factors recommended for use in scaling 

short-period response spectral acceleration for building-code 

and engineering applications have a weaker dependence on 

cyclical strain than that predicted for PHA and, therefore, 

short-period spectral acceleration, by empirical attenuation 

relationships. Borcherdt (1 994) described his recommended 

site factors as being based directly on Lorna Prieta strongmotion recordings for low-amplitude rock accelerations and 

on theoretical dynamic site-response analyses for higher amplitude rock accelerations. As demonstrated in table 6, the 

use of Loma Prieta observations to determine the low-strain 

amplification factors has resulted in factors that are consistent with a large number of other empirical observations. 

However, the results in table 7 indicate that, although the 

trend with amplitude is similar, the amplification factors recommended for building-code and engineering applications 

for hard-rock accelerations of 0.1 g and larger are consistently higher than those predicted from empirical attenuation relationships, This latter observation is independent of 

the assumption of whether a generic rock site is classified as 

either soft rock or hard rock. 

It is possible that the above discrepancy is the result of an 

intentional built-in conservatism in the recommended amplification factors, a bias in the theoretical site-response 

analyses that were used to calculate the high-strain dependence of these factors, and/or a bias in the way that the empiri- 
D64 PERFORMANCE OF THE BUILT ENVIRONMENT 

Table 7.-Summary of amplitude-dependent site-amplification factors for PHA and short-period response spectral acceleration for 

alluvium (SC-111, SD) relative to soft rock (SC-11, SC)' 

[PHA, mean peak horizontal acceleration; g, acceleration of gravity; MW, moment magnitude; -, no value] 

Reference PHA on hard rock (SC-Ib, SB) 

Empirical attenuation relationships (MW 6.5, 7.5) 

Abrahamson and Silva (1997) 1.15, 1.15 1.02, 1.02 0.89,0.89 

Boore and others (1997) 1.31, 1.31 - - 

Campbell and Bozorgnia (1994b), Campbell (1997) 1.21, 1.34 1.12, 1.23 1.01, 1.1 1 

Idriss (1991, 1993) 1.17, 1.17 1-06, 1.05 0.96,0.95 

Sadigh and others (1997) 1.15, 1.34 1.02, 1.15 0.91,0.99 

MEDIAN 1.20,1.26 1.05,l.ll 0.94,0.98 

MEDIAN (all MW) 1.23 1.08 0.96 

68% CONFIDENCE INTERVAL (all MW) 1.15-1.31 1.01-1.16 0.89-1.04 

Building-code and engineering applications 

Borcherdt (1994), Borcherdt and Glassmoyer (1994) 1.23 1.23 1.17 1 .OO 0.90 

Proposed 1997 UBC (International Conference of 1.332 1.273 1.17 1.09 1 .OO 

Building Officials, 1996) 

1Definition of site-classification criteria is given in table 1. 

2Corresponds to a hard-rock PHA of 0.075 g. 

3Corresponds to average of hard-rock PHA of 0.075 g and 0.15 g. 

cal attenuation relationships were developed. The possibility of the last consideration is demonstrated by the empirical predictions of Campbell and Bozorgnia (1994b) and 

Campbell (1997). These relationships were developed specifically to accommodate an independent scaling of site amplification with distance, thereby minimizing any bias in 

these factors imposed by arbitrary constraints. The results 

in table 7 show that the amplification factors predicted by 

these relationships are greater than those predicted by the 

other attenuation relationships, but generally consistent with 

those recommended for building-code and engineering applications. 

MODEL VERIFICATION 

The validity of the ground-motion models developed in 

this study was tested by comparing 37 estimates of PHA, 

PHV, and MMI derived from equations (2) through (4) with 

values observed on alluvium at sites located within about 

250 km of the seismogenic rupture zone (table 8). These sites 

were specifically selected to sample a wide range of distances and azimuths. A few sites located on Bay mud were 

included in the comparison to demonstrate the difference 

between expected and observed ground motions on soft soils 

located along the margins of San Francisco Bay. Because 

the MMI-PGA relationship predicts the MMI contour for a 

given value of PHA, only the integer part of the calculated 

MMI should be compared with the observed value of MMI. 

For example, calculated intensities of 6.3 and 6.7 are comparable to an observed MMI of VI (6). To make this comparison easier in table 8, the value of the calculated MMI 

that should be compared to the observed MMI is given in 

parentheses after the calculated value. 

The following adjustments were made to the predicted 

ground motions for Bay mud sites located in San Francisco 

and Redwood City to account for the presence of soft soils. 

Estimates of PHA from equation (2) were adjusted by the 

relative difference between peak accelerations recorded on 

soft soils (SC-IV, SE) and those recorded on alluvium (SC111, S,,) based on charts given by Seed and Idriss (1982) and 

Idriss (1990). Estimates of PHV from equation (4) were increased by a factor of 1.7 based on site-amplification factors 

developed by Campbell (1992). Estimates of MMI from equation (3) were increased by one intensity unit based on the 

expected difference in intensity between firm soils (GroundShaking Unit L) and soft soils (Ground-Shaking Unit J) recommended by Evernden and Thomson (1988). As discussed 

below, predictions at several other localities were also adjusted to reflect the unusually large ground motions and intensities observed at these locations during the earthquake. 

Strong-motion recordings obtained during the earthquake 

clearly showed that sites located on Bay mud had, on average, significantly higher ground motions than those located 

on alluvium or rock (see Boore and others, 1989; Campbell, 

199 1 ; Borcherdt and Glassmoyer, 1992, 1994). There were, 

however, some notable exceptions to these observations. For 

example, accelerograms recorded on Bay mud at Foster City 

and near the Dumbarton Bridge had peak accelerations similar to those recorded on nearby alluvial sites. 
EMPIRICAL ANALYSIS 

Table 8.-Comparison of predicted and observed values of MMI, PHA, and PHV 

[MMI, Modified Mercalli intensity; PHA, mean peak horizontal acceleration; PHV, mean peak horizontal velocity; Rs, closest 

distance to seismogenic rupture zone, Pred., predicted; Obs., observed or recorded; -, no value reported] 

Location 4 MMI PHA (g) PHV (cmls) 

Hollister ..................................... Oakland (Bay mud, Merritt Sand)------- 

Oakland (alluvium) ........................ San Francisco (Bay mud) ----------------- 

San Francisco (alluvium) ----------------- Gilroy----- ................................... Morgan Hill ---------- -- ---- ----- ----- - ------ San Jose --- ----------- ---- ------------ - ------ Palo Alto -------------- - -------------- - ------ 

'value in parentheses indicates number to be compared with observed value. 

There were also several sandy sites that exhibited amplifications comparable to those on Bay mud. One such site-a 

two-story office building in downtown Oakland-is founded 

on Merritt Sand, a Pleistocene deposit of saturated dune sand 

that overlies older bay sediments. These observations together 

with those noted above have led some investigators (see 

Shakal and others, 1990; Hanks and Brady, 199 1) to suggest 

that the response of the entire soil column above bedrock, 

not simply the response of Bay mud and other surficial soft 

sediments, was responsible for the observed amplification at 

many of these sites. As discussed previously, several investigators have suggested that critical reflections from within 

the crust and from the Moho were at least partly responsible 

for the high ground motions observed in San Francisco and 

Oakland. However, as discussed below, these critical reflections cannot totally explain the unusual pattern of ground 

motion and damage observed in parts of San Francisco and 

Oakland during the mainshock. 
D66 PERFORMANCE OF THE BUILT ENVIRONMENT 

MMI observations by Plafker and Galloway (1989) and 

E.V. Leyendecker (oral commun., 1992) indicate that damage on firm soils on the San Francisco Peninsula north of 

San Bruno were generally consistent with an MMI of VI, 

and that areas of saturated fill and Bay mud along the northern and eastern margins of San Francisco were consistent 

with an MMI of VII with isolated pockets of VIII and IX. 

The areas of higher intensity coincide with areas of relatively high ground motions observed during postearthquake 

seismological investigations (Boatwright and others, 199la, 

b; Seekins and Boatwright, 1994). In contrast, Stover and 

others (1990) assigned a uniform MMI of VII to most of the 

San Francisco Peninsula. E.V. Leyendecker (oral commun., 

1992) indicates that this assessment by Stover and others 

was somewhat conservative and does not reflect the distribution of damage observed from his field reconnaissance. 

Recorded ground motions in downtown Oakland on Merritt 

Sand were nearly as large as those recorded at nearby sites 

on Bay mud at the Oakland Harbor Wharf and at Emeryville. 

Yet, other nearby sites in east Oakland and Berkeley had 

ground motions significantly smaller than these. Damage in 

downtown Oakland and along the eastern margin of San Francisco Bay where saturated artificial fill and Bay mud predominate was found to be generally consistent with an MMI 

of VII with isolated pockets of VIII and IX (Plafker and 

Galloway, 1989; Stover and others, 1990; E.V. Leyendecker, 

oral commun., 1992). These observations are consistent with 

the high ground motions and intensities observed on similar 

sites in San Francisco. 

The lower ground motions recorded in Berkeley appear to 

be inconsistent with both the MMI of VII generally assigned 

to this area by Stover and others (1990) and the MMI of VI 

assigned to the area further north near Richmond, where somewhat higher ground motions were recorded. After reviewing 

his notes, E.V. Leyendecker (oral cornmun., 1992) suggested 

that, in his opinion, the area north of downtown Oakland should 

have been assigned an MMI of VI, more consistent with ground 

motions and damage observed in east Oakland and Richmond. 

Based on the above observations, we adjusted the predicted 

ground motions for sites located on saturated artificial fill 

and Bay mud, including those sites located along the northern and northeastern margins of the San Francisco Peninsula, and for sites located on Merritt Sand in downtown Oakland and near Alameda (south of Oakland) to be consistent 

with the increased ground shaking expected on soft soil. 

The region around the city of Hollister was another area 

that exhibited unusually high ground motions during the 

earthquake. Stover and others (1990) assigned an MMI of 

VIII to this area, whereas damage at comparable distances 

in this same general vicinity was found to be more consistent with an MMI of VII or less. Steidl and others (1991) and 

Wald and others (1991) were unable to account for these 

larger than expected ground motions in their seismological 

models of the main shock and suggested that site effects may 

have been responsible for the higher ground motions in this 

area. Because of the strong empirical and theoretical evidence that indicated unusually high site amplification in this 

area, the predicted ground motions for sites located in the 

Hollister region were adjusted to be consistent with the increased ground shaking expected on soft soil. This adjustment produces estimates of PHA, PHV, and MMI that are 

consistent with those observed in the area (table 8). 

In conclusion, the comparison shown in table 8 indicates 

that the ground-motion models developed in this study accurately predict both the peak ground-motion parameters and 

the Modified Mercalli intensities observed throughout the 

central California region during the earthquake over a wide 

range of azimuths and distances. 

ACKNOWLEDGMENTS 

Dave Hampson, formerly with the U.S. Geological Survey and now with EQE International, Inc., assisted in developing the geologic descriptions of the strong-motion recording sites used in this study. Glen Reagor of the U.S. Geological Survey provided a digital database of MMI observations collected by the USGS. Klaus Jacob of Larnont-Doherty 

Geological Observatory, Roger Borcherdt of the U.S. Geological Survey, and several anonymous reviewers provided 

insightful comments that significantly improved the manuscript. The study was initiated at the U.S. Geological Survey 

and was completed at Dames & Moore and EQE International. The author would like to thank these organizations 

for their support. 

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EMPIRICAL ANALYSIS D67 

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1987, Predicting strong ground motion in Utah, in Gori, 

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1989, Empirical prediction of near-source ground motion for 

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1990, Empirical prediction of near-source soil and soft-rock 

ground motion for the Diablo Canyon Power Plant site, San 

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Dames & Moore. 

199 1, An empirical analysis of peak horizontal acceleration 

for the Loma Prieta, California, earthquake of 18 October 1989: 

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1838-1858. 

1992, Recommended models for predicting strong ground motion and MMI in the San Francisco Bay area: Evergreen, Colorado, 

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1, p. 154-179. 

Campbell, K. W. and Bozorgnia, Y., 1994a, Empirical analysis of 

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to accompany tapes LOMAPRIETA89-IG (phase 1 data) and 

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199 1 b, Plots to accompany tapes LOMAPRIETA89-IB 1, - 

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(phase 1 data) and LOMAPRIETA89-L (phases 2 & 3 data), 

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important ground motion parameters associated with large and 

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Fumal, T.E., 199 1, A compilation of the geology and measured and 

estimated shear-wave velocity profiles at strong-motion stations that recorded the Loma Prieta, California, earthquake: 

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ground motion, and damage in Oakland, Treasure Island, and 

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report to Governor George Deukrnejian from The Governor's 

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revisions to the 1994 editions of the uniform codes (submittals for 1996): Whittier, California. 
PERFORMANCE OF THE BUILT ENVIRONMENT 

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for the 1989 M 6.9 Loma Prieta earthquake including effects of 

source, path, and site: Earthquake Spectra, v. 9, p. 251-287. 

Seed, H.B., and Idriss, I.M., 1982, Ground motions and soil liquefaction during earthquakes, in Agbabian, M.S., ed., Engineering Monographs on Earthquake Criteria, Structural Design, 

and Strong Motion Records: Berkeley, California, Earthquake 

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Seekins, L.C., and Boatwright, J., 1994, Ground motion amplification, geology, and damage from the 1989 Loma Prieta earthquake in the City of San Francisco: Bulletin of the Seismological Society of America, v. 84, p. 18-30. 

Shakal, A., Huang, M., Reichle, M., Ventura, C., Cao, T., Sherburne, 

R., Savage, M., Darragh, R., and Petersen, C., 1989, CSMIP 

strong-motion records from the Santa Cruz Mountains (Loma 

Prieta), California, earthquake of 17 October 1989: Sacramento, California, Division of Mines and Geology, Strong 

Motion Instrumentation Program, Report no. OSMS 89-06. 

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Strong ground shaking from the Loma Prieta earthquake of 17 

October 1989, and its relation to near surface geology in the 

Oakland area, in McNutt, S.R., and Sydnor, R.H., eds., The 

Loma Prieta (Santa Cruz Mountains) California, earthquake 

of 17 October 1989: Sacramento, California, Division of Mines 

and Geology, Special Report 104. 

Somerville, P., and Yoshimura, J., 1990, The influence of critical 

MOHO reflections on strong ground motions recorded in San 

Francisco and Oakland during the 1989 Loma Prieta earthquake: Geophysical Research Letters, v. 17, p. 1203- 1206. 

Somerville, P.G., Smith, N.F., and Graves, R.W., 1994, Effect of 

critical reflections from the Moho on the attenuation of strong 

ground motion, in Borcherdt, R.D., ed., The Loma Prieta, 

California, earthquake of October 17, 1989-strong ground 

motion: U.S. Geological Survey Professional Paper 155 1-A, 

p. 67-75. 

Steidl, J.H., Archuleta, R.J., and Hartzell, S.H., 1991, Rupture history of the 1989 Lorna Prieta, California, earthquake: Bulletin 

of the Seismological Society of America, v. 8 1, p. 1573-1602. 

Stevens, J.L., and Day, S.M., 1994, Simulation of strong ground 

motion, in Borcherdt, R.D., ed., The Loma Prieta, California, 

earthquake of October 17,1989-strong ground motion: U.S. 

Geological Survey Professional Paper 1551-A, p. 53-60. 

Stover, C.W., Reagor, G.R., Baldwin, F.W., and Brewer, L.R., 1990,' 

Preliminary isoseismal map for the Santa Cruz (Loma Prieta), 

California, earthquake of October 18, 1989 UTC: U.S. Geological Survey Open-File Report 90- 18. 

Trifunac, M.D., 1976, A note on the range of peak amplitudes of 

recorded accelerations, velocities and displacements with respect to the Modified Mercalli intensity scale: Earthquake 

Notes, v. 47, no. 1, p. 9-24. 

U.S. Geological Survey Staff, 1990. The Loma Prieta, California, 

earthquake-an anticipated event: Science, v. 247, p. 286-293. 

Wald, D.J., Helmberger, D.V., and Heaton, T.H., 1991, Rupture 

model of the 1989 Loma Prieta earthquake from the inversion 

of strong-motion and broadband teleseismic data: Bulletin of 

the Seismological Society of America, v. 8 1, p. 1540- 1572. 
THE LOMA PRIETA, CALIFORNIA, EARTHQUAKE OF OCTOBER 17,1989: 

PERFORMANCE OF THE BUILT ENVIRONMENT 

EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

ATTENUATION OF VERTICAL AND HORIZONTAL RESPONSE SPECTRA 

OF THE LOMA PRIETA EARTHQUAKE 

By Yousef Bozorgnia, Exponent Failure Analysis Associates, Menlo Park, Calif.; and 

Mansour Niazi, Berkeley Geophysical Consultants, Berkeley, Calif. 

CONTENTS 

Page 

D69 

69 

70 

70 

72 

76 

76 

78 

79 

79 

79 

ABSTRACT 

Ground motion attenuation relationships of vertical and 

horizontal response spectra for the Loma Prieta earthquake 

are developed through analyses of the ground motion at 53 

sites within a 100 km radius of the source. The analyses are 

performed on the spectral ordinates for 16 incremental periods ranging from 0.05 to 2.0 seconds. The response spectra 

are modeled empirically for site conditions characterized by 

rock and by stiff soil. Data analysis is performed by the application of a nonlinear multivariate regression procedure 

allowing for distance and site factor as independent variables. Variation of the vertical-to-horizontal (VIH) spectral 

ratios with wave frequency and distance shows the same 

behavior as observed previously in the widely separated geographic regions of northeastern Taiwan and east-central Iran. 

The predicted ratios at sites underlain by stiff soil are generally higher than the commonly used value of two-thirds at 

high frequencies (>5 Hz) in the near-source region (Re30 

km) but reduce to one-half or less at longer periods and farther distances. This behavior is also observed at rock sites; 

however, it is somewhat less pronounced. 

With a faster attenuation of spectral ordinates at higher 

frequencies, the shape of the response spectrum is found to 

change with distance. As expected, the spectral attenuation 

with distance is generally higher for the vertical spectrum 

than for the horizontal spectrum. The difference is particularly significant at the higher frequency end of the spectrum. 

Site amplification factors for stiff soil with respect to rock 

vary between 1.17 and 1.72 for the horizontal spectrum and 

between 1 .O1 and 1.8 1 for the vertical spectrum. Spectral 

amplifications at four sites underlain by soft soil and artificial fill are also evaluated. This is done by a comparison of 

the observed spectra with those predicted for rock sites at 

corresponding distances. As expected, the resulting amplification factors at soft-soil sites show significant increase relative to those at sites underlain by rock. 

INTRODUCTION 

The earthquake was the strongest event to occur on this 

segment of the San Andreas fault since the great San Francisco earthquake of 1906. It caused loss of life and considerable damage in the region, including the collapse of the Cypress section of Interstate Highway 880. The details of source 

geometry and extent of damage are given in U.S. Geological 

Survey (1989a), Earthquake Engineer Research Institute 

(1990), and Housner (1990). The epicentral region was well 

instrumented, including over a hundred accelerographs which 

provided an important suite of high-quality triaxial 

accelerograms. The behavior of the peak ground acceleration (PGA) has previously been studied by Boore and others 

(1989), Campbell (1991), and Niazi and others (1992). Despite some differences on the interpretation of the results, 

these studies suggest that the observed horizontal PGA for 

this earthquake exceeded previous model predictions. 

The purpose of this study is to extend the analysis of PGA 

to the response spectra and to assess the intensity of vertical 

ground motion in the frequency range of engineering interest. Both vertical and horizontal response spectra are analyzed, and the results are given for sites characterized by 

rock and by stiff soil. 
D70 EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

DATA SOURCE 

The database consists of the pseudovelocity response spectra (PSV) calculated at 5 percent of critical damping at 53 

sites within 100 km distance of the source as shown in figure 

1. Site geology at 22 sites is characterized as rock, 27 as stiff 

soil (alluvial sites), and 4 as soft soil (Bay mud and artificial 

fill). The majority of the stations are maintained by the Strong 

Motion Instrumentation Program (SMIP) project of the California Department of Conservation (Shakal and others, 1989; 

Huang and others, 1990), and by the U.S. Geological Survey Strong Motion Program (U.S. Geological Survey, 1989b). 

The spectral ordinates used in this study have been calculated and published by these agencies from bandpass filtered 

accelerograms. 

The analysis is performed at 16 incremental periods ranging between 0.05 and 2.0 s. Vertical PSV's are directly used 

in the analysis. For the horizontal PSV's , the arithmetic mean 

of the ordinates derived for the two horizontal components 

are entered. This is consistent with the previous studies by 

Niazi and Bozorgnia (1992a) and Campbell (1993). In total, 

nearly 2,500 spectral ordinates are analyzed. Table 1 gives a 

list of recording stations used in this study and their site geology. The distances are from the vertical projection of the 

rupture on the Earth's surface and have been discussed by 

Boore and others (1989) and Niazi and others (1992). The 

vertical and horizontal spectral ordinates at periods 0.1 and 

0.5 s are also given in table 1. 

DATA ANALYSIS 

Data analysis was performed by the application of a 

multiregression procedure (SAS, 1985) with distance and 

PACIFIC 

OCEAN 

A Rock 

* Stiff Soil 

X Soft Soil 

Projected Rupture Area I 

Figure 1 .-Locations of 53 recording sites considered in this study. The shaded area is the surface projection of the 

rupture plane (Boore and others, 1989; Niazi and others, 1992). 
ATTENUATION OF VERTICAL AND HORIZONTAL RESPONSE SPECTRA OF THE LOMA PRIETA EARTHQUAKE D7 1 

Table 1 .-Recording stations used in this study 

HOR. PSV 

at period 

0.5 s 

(ids) 

Recording 

stations 

Distance 

(km) 

Corralitos 

Capitola 

Gilroy#l 

Gilroy G C 

Saratoga 

Gilroy#2 

Gilroy Cm Bldg 

Santa Cruz 

San Js. St.Te. 

Gilroy#3 

Gilroy#4 

Gilroy#6 

Hollister W.H 1 28.3 1 0 1 1.750 1 4.640 1 2.005 

ROCK (**) 

0.1 

8.6 

10.5 

10.9 

Anderson D.ds 

Coyote L ds 

Gilroy#7 MR 

Hollister AP 

Agnews SH 

Sunnyv. Colt. 

Hollister CH 

11.7 

12.1 

1 2.3 

12.5 

13.2 

14.0 

15.8 

19.0 

Vertical PSV 

at period 

0.1 s 

(inls) 

1 

0 

1 

0 

20.0 

21.7 

24.3 

25.4 

27 .O 

27.5 

27.8 

, , I 

0 

0 

0 

1 

0 

0 

0 

1 

Hollister S&P 1 28.3 

Halls Valley 

SAGO Vault 

Vertical PSV 

at period 

0.5 s 

(ink) 

5.230 

8.880 

1.750 

2.250 

0 

0 

0 

0 

0 

0 - 

0 

Salinas 

Milpitas 2s 

Sago S 

Stanford SLAC 

Calveras Res. 

Men10 Park VA 

Woodside 

Fremont Msj 

Monterey CH 

U Crystal 10 1 46.6 1 1 1 .298 I 3.060 I .700 

HOR. PSV 

at period 

0.1 s 

(inls) 

4.720 

3.070 

2.8 10 

8.980 

4.950 

3.640 

2.250 

1.370 

0 1 2.610 1 4.920 

29.3 

29.9 

t - - , I 

I Foster City 1 47.3 * 1.650 1 4.500 1 1 395 

13.600 

6.610 

6.670 

4.770 

1.629 

1.490 

1.550 

2.256 

1.240 

1.742 ---- 

2.058 

1.995 

31.4 

31.4 

34.1 

35.0 

36.1 

36.5 

38.7 

42.0 

42.7 

U Crystal #7 1 46.5 

Hayward JMS 1 53.6 1 0 1 1.250 I 1.790 I 1.565 

4.325 

4.675 

5.965 

4.485 

7.740 

5.320 

5.270 

3.880 

6.040 

7.000 

3.290 

3.650 

0 

1 

(*) Sites specified by an asterisk (*) are on soft soil. 

(**) Site parameter ROCK is 1 for rock sites and 0 for stiff-soil sites. 

4.030 

3.145 

2.130 

4.730 

4.545 

5.980 

2.545 

1.825 

6.276 

4.960 

1.630 

4.000 

5.130 

3.407 

4.000 

0 

0 

1 

1 

0 

0 

0 

0 

1 

1 

2.474 

1.670 

2.5 10 

2.042 

1.510 

2.029 

1.492 

,843 

.612 - 

1.520 

1.290 

.467 

1.885 

1.026 

2.696 

388 

1.450 

.434 

.68 1 1 3.070 

6.980 

1.796 

.977 

962 

-570 

2.670 

2.980 

3.610 

6.043 

1.603 

2.701 

2.000 

4.460 

1.480 

,901 

1.035 

.579 

2.099 

.900 

1.719 

.597 

1.1 85 

.783 
D72 EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

Table 1 .-Continued 

Recording Distance 

stations 

~ayward CSUHFY 76.0 - 

Hayward BART 57.1 

SF0 63.2 

S SF Si.Pt. 67.6 

SF Dim Hights 75.9 

Piedmont Jr.H 77.2 

SF Rincon Hill 78.5 

Oaklnd 0 Har W 78.8 

Yerba Buena I 79.5 

SF Pac Hights 80.5 

SF Tel Hill 80.9 

Tresure I 81.7 

Presidio 81.9 

LBL #l 83.2 

SF Cliff House 83.2 

Pt. Bonita 87.5 

Richmond CH 92.0 

ROCK (**) Vertical PSV Vertical PSV HOR. PSV HOR- PSV 

at period at period at period at period 

0.1 s 0.5 s 0.1 s 0.5 s 

(inls) (inls) (inls) (inls) 

1 -496 2.590 .83 1 3 -740 

0 -972 3.640 2.240 5.970 

* -474 3 -750 1.940 21.150 

(*) Sites specified by an asterisk (*) are on soft soil. 

(**) Site parameter ROCK is 1 for rock sites and 0 for stiff-soil sites. 

site geology as independent variables. Hence, as seen in equation (I), the site parameter "s" would allow an adjustment 

for site geology of PSV for stiff soil relative to rock. 

In this equation R is distance in kilometers, PSV is 

pseudovelocity spectral ordinate in incheslsecond, ROCK is 

a site variable which has a value of 1 for rock and 0 for stiff 

soil, and a, c, d, and s are regression parameters. The input 

to the regression is confined only to the information at 49 

sites characterized by either rock or stiff soil. The four softsoil sites are used later for the evaluation of the amplification due to soft soil material. In equation (1) e represents 

random error with zero mean and o2 variance. 

A nonlinear regression procedure (SAS, 1985) is applied 

to the spectral ordinates at each of the 16 frequencies. A subjective weighting method is introduced to compensate for 

the concentration of the observations within specific distance 

bins. Because the employed methodology and the set up is 

similar to the one used in the study of PGA behavior by Niazi 

and others (19921, for further detail the reader is referred to 

that study. 

DISTANCE-DEPENDENT 

RESPONSE SPECTRA 

The numerical results for the regression parameters derived for the horizontal and vertical components are given 

in tables 2 and 3, respectively, as functions of period. 

Figure 2 shows the attenuation of the 5 percent damped 

vertical and horizontal response spectra at a period of 0.1 

s for sites with stiff soil geology. The observed spectral 

ordinates for stiff soil sites are also marked in figure 2. 

Figures 3 and 4 show the attenuation of spectra at 0.5 and 

1.0 s, respectively. It is evident from figures 2 through 4 

that the magnitude of the vertical and horizontal spectral 

ordinates are comparable at the short period (0.1 s), while 

at the longer periods the horizontal spectra exceed those 

for vertical motion. This is due to the fact that the vertical 

ground motion is in general enriched in the higher frequency waves relative to the horizontal component. Thus, 

at the short period end of the spectrum, the vertical motion is pronounced. At longer periods, however, the vertical motion does not contain enough energy to strongly 

excite a long-period oscillator; therefore, as shown in fig- 

EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

Figure 2.-Attenuation of 5 percent damped vertical (A) and horizontal (B) response spectra at a period of 0.1 s for stiff soil sites. Solid curve 

is for the median spectrum. One standard deviation band is shown by the dashed curves. The observed spectral ordinates are also marked. 

Figure 3.-Attenuation of 5 percent damped vertical (A) and horizontal (5) spectra at a period of 0.5 s for stiff soil sites. Solid curve is for 

the median spectrum. One standard deviation band is shown by the dashed curves. The observed spectral ordinates are also marked. 

cates that scaling of a standard spectral shape with a single 

parameter, such as PGA, is not justified. This was also noted 

by Krawinkler and others (1991) regarding the horizontal 

spectrum. 

Comparison of figures 5Aand 5B shows that the spectral 

ordinates at stiff-soil sites are generally larger than those at 

rock sites. This is also true for vertical spectra (fig. 6). except, by different factors. It is also evident from tables 2 and 

3 where the site parameter "s" for soil sites is negative for 

both horizontal and vertical spectra. The resulting amplification factor at stiff-soil sites relative to rock varies between 

1.17 and 1.72 for the horizontal spectrum and between 1 .O1 

and 1.8 1 for the vertical spectrum. Application of the regression procedure to equation (1) at discrete periods results in a 

site parameter "s" as a function of the wave period. The 

model adopted in this study, however, does not allow evalu- 
ATTENUATION OF VERTICAL AND HORIZONTAL RESPONSE SPECTRA OF THE LOMA PRIETA EARTHQUAKE D75 

Figure 4.-Attenuation of 5 

percent damped vertical (A) 

and horizontal (5) spectra at 

period 1.0 s for stiff soil 

sites. Solid curve is for the 

median spectrum. One standard deviation band is 

shown by the dashed curves. 

The observed spectral ordinates are also marked. 

ROCK SITE 

Figure 5.-Attenuation 

of 5 percent damped 

spectra: Median horizontal response spectra 

at distances R=lO, 20, 

40,60 km from the surface projection of the 

rupture plane for (A) stiff 

soil and (B) rock sites. 

STIFF SOIL 

/ 

Figure 6.-Attenuation 

of 5 percent damped 

spectra: Median vertical 

response spectra at distances R=10,20,40,60 

km from the surface projection of the rupture 

plane for (A) stiff soil 

and (B) rock sites. 

STIFF SOIL A - ROCK SITE B 
D76 EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

ation of distance dependence of the parameter "s,"unless, as 

explained below, independent analyses are performed on the 

subsets of data collected at rock and soil sites, separately. It 

is of interest to note that for the vertical spectrum the site 

amplification factor "s" generally tends to decrease with 

period, whereas for the horizontal spectrum the trend is reversed, especially for periods longer than about 0.3 s (see 

tables 2 and 3). 

UNIFORM GEOLOGY MODELS 

The regression coefficients of tables 2 and 3, and the resulting empirical spectra of figures 5 and 6 are all based on 

the analyses of the entire data for both stiff soil and rock 

sites combined. Next, the data files are split into two subsets, one containing stiff-soil data and the other rock data. 

By doing so we search for possible bias in the predictions at 

stiff-soil sites which may be introduced by the observations 

at rock sites and vice versa. As already explained, there are 

22 rock sites and 27 sites characterized by stiff soil in the 

data. Application of the regression procedure given in equation (I), without the soil classification term, to the two data 

subsets, results in four sets of independent regression coefficients which may somewhat differ from those listed in tables 

2 and 3. The resulting empirical models of horizontal and 

vertical pseudoacceleration spectra (PSA) obtained from the 

two procedures are compared in figures 7 and 8, respectively. 

Figure 7A compares the horizontal acceleration spectra for 

the subset of data observed at stiff-soil sites (dashed curves) 

R=10 

SET 

and for the combined data (solid curves) at 10,20,40 and 60 

km. Figure 75 shows the same comparison at rock sites. 

Corresponding plots for the vertical component are shown 

in figure 8, in which again figure 8A compares the predictions of stiff-soil observations with those of the combined 

data and figure 8B compares those of rock observations and 

the entire data. It is noted that the largest separation between 

the dashed and solid curves occurs at near-source distances 

of up to 10 krn, where very few observations exist, especially for rock sites. To provide a scale for the comparison, 

in each example we have also plotted the mean plus sigma 

(84.1 percent probability) of predictions at 10 km distance 

for the combined set. In comparison with one sigma, even 

the largest observed differences between the prediction results are thus relatively small and the models based on the 

entire data are confirmed. 

VERTICAL-TO-HORIZONTAL 

SPECTRAL RATIOS 

Using the regression parameters of tables 2 and 3, vertical-to-horizontal ratios (VIH) of the predicted response spectra at four given distances of 10, 20, 40 and 60 km are 

derived and shown in figure 9. It is clear from figure 9 that 

V/H spectral ratio is distance-as well as frequency-dependent. The highest ratios are observed for the high-frequency motion in the near-source region (Niazi and 

Bozorgnia, 1992b). At more distant sites, the spectral ratios 

decrease apparently due to a faster rate of attenuation for the 

Q 0.80 - 

or 0.60 - 

1 

0.40 - 

-^ 0.20 - E& - LiJU so. 

R=10 

SET 

Figure 7.Ã‘Compariso of the results of two regression analyses on the spectral ordinates: median horizontal response spectra at distances 

R=10,20,40,60 km for (A) stiff soil and (B) rock sites. Solid curves are for the application of equation (1) to the combined soil and rock 

data and assigning proper value for variable "ROCK." Dashed curves are for the analysis of (A) soil data only and (6) rock data only. The 

84th percentile spectra for R=10 km of the combined analysis are also shown. 
ATTENUATION OF VERTICAL AND HORIZONTAL RESPONSE SPECTRA OF THE LOMA PRIETA EARTHQUAKE D77 

high-frequency-rich vertical motion as compared to the low- and Bozorgnia, 1992a). Spectral ratios generally exceed the 

frequency horizontal component. commonly used value of two-thirds at periods shorter than 

The observed variation with frequency and distance at stiff- 0.2 s, particularly at near-source distances of R < 30 km. 

soil sites (fig. 9A) shows similar behavior as those reported The ratios, however, fall below one-half at longer periods 

previously at stiff-soil sites in widely separated geographic for all distances; that is, the use of two-thirds factor would 

regions of northeastern Taiwan and east-central Iran (Niazi likely result in overconservatism at long periods. Spectral 

Ft f,,' '\, 

'' fa' + 

t ,\ /, ; ; 1 \ , , /M+SIGMA, R=10 

,' 1 , COMBINED SET 

-0 u0- I, 

"&' \ 

ft' I I 

$1-M+SIGMA, R= 1 o 

COMBINED SET 

R=10 

R=20 

R=40 

R=60 

Figure 8.-Comparison of the results of two regression analyses on the spectral ordinates: Median vertical response spectra at distances 

R=10,20,40,60 km for (A) stiff soil and (B) rock sites. Solid curves are for the application of equation (1) to the combined soil and rock 

data and assigning proper value for variable "ROCK." Dashed curves are for the analysis of (A) soil data only and (B) rock data only. The 

84th percentile spectrum for R=10 km of the combined analysis is also shown. 

Figure 9.-Ratio of predicted vertical to horizontal spectral ordinates at distances R=10,20,40, and 60 km from the source at (A) stiff soil 

and (B) rock sites. Regression was performed on the combined soil and rock data set. The ratios of vertical to horizontal PGA's are also 

given (Niazi and others, 1992). 
D78 EARTH STRUCTURES AND ENGINEERING CHARACTERIZATION OF GROUND MOTION 

ratios for different distances converge to a common point at 

a period of about 0.3 s, where the ratio becomes independent 

of distance at a value of 0.4. 

The same behavior is also observed at rock sites (fig. 9B), 

however, not as pronounced as for soil. The V/H spectral 

ratios at stiff-soil sites are larger than those at rock sites for 

periods shorter than 0.4 s, and slightly less for longer periods. The predicted values of the V/H ratio of PGA for this 

earthquake are also listed in the insets of figures 9A and 9B 

for the corresponding site geology. These values are based 

on the empirical predictions of Niazi and others (1992). 

We offer the following preliminary explanation for the 

observed variation of the V/H spectral ratio. Since the vertical motion is enriched in higher frequencies, it will be attenuated at a faster pace. The behavior of V/H spectral ratio 

is due to the combination of two factors. The first factor is 

the differential attenuation of higher frequency P waves (predominantly vertical) relative to shear waves. The second factor is the ray geometry, whereby at short distances the highfrequency P waves have a larger vertical component than 

shear waves, owing to a near-vertical incidence and alignment of particle motion with the direction of propagation. 

The VIH spectral behavior observed in several regions, 

over a wide range of peak accelerations, therefore appears to 

be real. The existence of deep sedimentary cover (stiff soil) 

apparently accentuates this behavior. Whether the nonlinear 

response of the soil has an influence is not clear. Observations by Darragh and Shakal(1991) of the main shock and 

several aftershocks at the Treasure Island site clearly demonstrate that horizontal response at sites underlain by soft 

sedimentary material and artificial fill is modified substantially by the nonlinear behavior of the soil. However, the same 

* SF0 

0 OAKLAND WHARF 0 TREASURE ISLAND A FOSTER CITY 9 

authors failed to detect a similar effect at the Gilroy#2 station with stiff-soil geology. 

SPECTRAL AMPLIFICATION OF 

SOFT SOIL 

In order to evaluate site amplification at stations underlain by soft soil, the observed response spectra at four sites 

with soft-soil geology are compared with the predictions of 

our empirical results at rock sites, computed at corresponding distances. These four sites are denoted by asterisk (*) in 

table 1. Figure 10 shows the result of this comparison at individual sites for both horizontal and vertical components. 

The observed site amplifications of horizontal spectra, as 

seen in figure 10A, show similar features at the four sites 

considered here. Spectral ratios show a gentle negative trend 

approximately up to the 0.2-s period. For the Oakland Wharf, 

Treasure Island, and Foster City sites, amplification peaks 

occur at 0.6-0.7-s and 1.5-s periods. For the San Francisco 

Airport (SFO), these peaks occur at shorter periods, implying a thinner soft-soil column. 

The peak horizontal spectral amplification varies between 

4.5 and 7 at these four sites. Darragh and Shakal(199 1) also 

found a peak Fourier spectral ratio of 4 for the strong phase 

of the horizontal motion at Treasure Island relative to the 

neighboring rock site at Yerba Buena Island. However, they 

also noted that the observed site response is systematically 

higher for smaller magnitude aftershocks. 

Site amplification for the vertical component (figure 105) 

is generally smaller than that of horizontal component. Except for the 1.5s peak at the Oakland Wharf with an ampli8'oo * SF0 

0 OAKLAND WHARF 6 

TREASURE ISLAND A FOSTER CITY 

Figure 10.-Ratio of observed spectra at soft-soil sites over predicted spectra at rock sites for (A) horizontal and (5) vertical ground 

motion. 
ATTENUATION OF VERTICAL AND HORIZONTAL RESPONSE SPECTRA OF THE LOMA PRIETA EARTHQUAKE D79 

fication factor of 3.8, peak amplifications occur predominantly at periods shorter than 0.3 s. As already reported by 

Niazi and others (1992), the vertical component at the Treasure Island site shows deamplification rather than amplification. 

DISCUSSION OF RESULTS 

Previous studies of a more extensive database recorded at 

SMART-1 array, Taiwan (Niazi and Bozorgnia, 1989a, 1989b, 

1992a), showed, for the first time, a strong dependence of 

VIH spectral ratios on both distance and wave frequency. 

Some of the suggestions emerging from that study are as 

follows: 

l.The shape of the response spectra for both vertical and 

horizontal components of the ground motion, and their ratio, is distance-as well as magnitude-dependent. 

2.The V/H spectral ratio at the high-frequency end of the 

spectrum shows a substantial increase relative to the standard two-thirds value. The increase is particularly significant at near-source distances. 

3.At low frequencies and far-field distances, the spectral 

ratios reduce significantly, such that the two-thirds ratio 

would become conservative. 

Most of these suggestions, with the exception of the magnitude-scaling effects (inapplicable to a single earthquake) 

also apply to the Loma Prieta observations. A study of the 

magnitude-scaling effect on this ratio requires simultaneous 

analysis of several earthquakes over a broad range of magnitude. 

Table 3 shows that in general the short-period end of the 

vertical response spectrum is characterized by higher values 

of the attenuation parameter d, site amplification parameter 

s, and to some extent the distance saturation parameter (depth 

term) c. However, the absolute value of d tends to decrease 

almost monotonically down to a value of 0.22 at 1 Hz. Below 1 Hz, the long-period noise may become significant. 

The correlation of attenuation with frequency is expected as 

a consequence of intrinsic attenuation and scattering. There 

may also be a slight possibility of a tradeoff between c and d 

parameters. For the horizontal motion, a similar trend is observed in the variation of the attenuation parameter d and the 

depth term c; however, vertical motion attenuates much faster 

at periods below 0.2 s than the horizontal motion (see tables 

2 and 3). The same differential attenuation between vertical 

and horizontal motions was also observed in the analysis of 

the SMART- 1 data below 0.2 s (Niazi and Bozorgnia, 1992a). 

However, the SMART- 1 data showed much higher attenuation at periods above 0.2 s than Loma Prieta for both vertical 

and horizontal components. Tables 2 and 3 show a tendency 

for the site-amplification factor to decrease with period for 

the vertical spectrum but to increase for the horizontal component, especially for periods longer than about 0.3 s. 

As recognized by previous investigators, the effects of the 

vertical ground motion on the response of structures may 

not always be neglected. For example, as Newmark and Hall 

(1982) indicated, columns and walls in compression, beams, 

and other horizontal elements are particularly vulnerable to 

the vertical component of ground motion. Other examples 

include the effects on the seismic response of multistory 

frames (Anderson and Bertero, 1973), concrete gravity dams 

(Chakrabarti and Chopra, 1973), and reinforced concrete 

highway bridges (Saadeghvaziri and Foutch, 1991). 

In the estimation of near-source vertical ground motion so 

far, reliance has been placed on the extrapolation of far-field 

information due to the scarcity of high-quality observations 

close to the source. Recent studies of SMART-1 data as well 

as observations of the 1978 Tabas, Iran, and the Loma Prieta 

earthquakes made it possible to quantify the range of distances and frequencies where the vertical ground motion is 

pronounced. 

ACKNOWLEDGMENTS 

The authors wish to thank the U.S. Geological Survey 

and California Department of Conservation for providing the 

accelerograms used in this study. Copyright release provided 

by John Wiley & Sons, Ltd, is also acknowledged. 

REFERENCES 

Anderson, J.W., and Bertero,V.V., 1973, Effects of gravity loads and 

vertical ground acceleration on the seismic response of multistory frames, in Proceedings of the Fifth World Conference on 

Earthquake Engineering, Rome, p. 2914-2923. 

Boore, D.M., Seekins, L., and Joyner,W.B., 1989, Peak accelerations 

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Campbell, K.W., 1991, An empirical analysis of peak horizontal acceleration for the Loma Prieta, California earthquake of October 18,1989: Bulletin of the Seismological Society of America, 

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1993, Empirical prediction of near-source ground motion from 

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Chakrabarti, P., and Chopra, A.K. 1973, Hydrodynamic pressures 

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Earthquake Engineering and Structural Dynamics, v. 1, p. 325-335. 

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1899. 

Earthquake Engineering Research Institute, 1990, Loma Prieta Earthquake Reconnaissance Report, Supp. Earthquake Spectra, v. 6. 

Housner, G.W., 1990, Competing against time; report to Governor G. 

Deukmejian, Board of Inquiry on the 1989 Loma Prieta earthquake: The State of California, Office of Planning and Research, 

Sacramento. 
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Huang, M.J., Cao, T.Q.,Vetter, U.R., and Shakal, A.F.,1990, Third 

interim set of CSMIP processed strong-motion records from the 

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17 October 1989: California Strong Motion Instrumentation 

Program, California Department of Conservation, Division of 

Mines and Geology, Report No. OSMS 90-05. 

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August 1989. 

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peak ground acceleration in Central California from observations of the October 17, 1989 Lorna Prieta earthquake: Earthquake Engineering and Structural Dynamics, v. 2 1, p. 493-507. 

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horizontal earthquake motions: Earthquake Engineering and 

Structural Dynamics, v. 20, p. 535-549. 

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N.C. 

Shakal, A., Huang, M., Reichle, M., Ventura, C., Cao, T., Sherburne, 

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strong-motion records from the Santa Cruz Mountains (Loma 

Prieta), California, earthquake of 17 October 1989, Report No. 

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A U.S. GOVERNMENT PRINTING OFFICE: 1998 - 773-051 / 20126 Region No. 8