United States Office of Research and EPA/600/R-00/020 

Environmental Protection Development January 2000 

Agency Washington, DC 20460 

Retrofitting Control Facilities For 

Wet-Weather Flow Treatment 

Research Report 
EPA#/600/R-00/020 

January 2000 

Retrofitting Control Facilities For 

Wet-Weather Flow Treatment 

by 

Peter E. Moffa, Howard M. Goebel, 

Daniel P. Davis and John J. LaGorga 

Moffa & Associates Consulting Engineers 

Syracuse, New York 13214 

In Association with 

Earth Tech, Inc. 

South Portland, Maine 04106 

Robert Pitt 

Environmental Engineer 

Birmingham, AL 35226 

Contract No. 8C-R056NTSX 

Project Officer 

Thomas P. O'Connor 

Technical Advisors
 

Richard Field & Mary K. Stinson
 

Urban Watershed Management Branch
 

Water Supply & Water Resources Division
 

National Risk Management Research Laboratory
 

Edison, NJ 08837
 

National Risk Management Research Laboratory
 

Office of Research and Development
 

U.S. Environmental Protection Agency
 

Cincinnati, Ohio 45268
 
Notice 

The information in this document has been funded by the United States Environmental 

Protection Agency under Contract No. 8C-R056-NTSX. Although it has been subjected 

to the Agency's peer and administrative review and has been approved for publication as 

an EPA document, it does not necessarily reflect the views of the Agency and no official 

endorsement should be inferred. Also, the mention of trade names or commercial 

products does not imply endorsement by the United States government. 

ii 
Foreword 

The U.S. Environmental Protection Agency is charged by Congress with protecting 

the Nation’s land, air, and water resources. Under a mandate of national environmental 

laws, the Agency strives to formulate and implement actions leading to a compatible 

balance between human activities and the ability of natural systems to support and 

nurture life. To meet this mandate, EPA’s research program is providing data and 

technical support for solving environmental problems today and building a science 

knowledge base necessary to manage our ecological resources wisely, understand how 

pollutants affect our health, and prevent or reduce environmental risks in the future. 

The National Risk Management Research Laboratory is the Agency’s center for 

investigation of technological and management approaches for reducing risk from threats 

to human health and the environment. The focus of the Laboratory’s research program is 

on methods for the prevention and control of pollution to air, land, water and subsurface 

resources; protection of water quality in public water systems; remediation of 

contaminated sites and ground water; and prevention and control of indoor air pollution. 

The goal of this research effort is to catalyze development and implementation of 

innovative, cost-effective, environmental technologies; develop scientific and engineering 

information needed by EPA to support regulatory and policy decisions; and provide 

technical support and information transfer to ensure effective implementation of 

environmental regulations and strategies. 

This publication has been produced as part of the Laboratory’s strategic long-term 

research plan. It is published and made available by EPA’s Office of Research and 

Development to assist the user community and to link researchers with their clients. 

E. Timothy Oppelt, Director
 

National Risk Management Research Laboratory
 

iii 
Abstract 

Available technologies were evaluated to demonstrate the technical feasibility and cost 

effectiveness of retrofitting existing facilities to handle wet-weather flow. Cost/benefit 

relationships were also compared to construction of new conventional control and 

treatment facilities. 

Desktop analyses of 13 separate retrofit examples were performed for 1) converting or 

retrofitting primary settling tanks with dissolved air flotation and lamellae and/or 

microsand-enhanced plate or tube settling units, 2) retrofitting existing wet-weather flow 

storage tanks to provide enhanced settling/ treatment and post-storm solids removal, 3) 

converting dry ponds to wet ponds for enhanced treatment, 4) retrofitting wet-weather 

flow storage tanks for dry-weather flow augmentation, 5) using storage for sanitary sewer 

overflow control , 6) retrofitting for industrial wastewater control in a combined sewer 

system, and 7) bringing outdated/abandoned treatment plants back online as wet-weather 

flow treatment facilities. 

This analysis demonstrated that retrofitting existing wet-weather flow facilities can be 

technically feasible in most cases and may be more cost effective than construction of 

new conventional control and treatment facilities. The feasibility and cost effectiveness of 

retrofitting was found to be a function of site-specific conditions and treatment 

requirements. Retrofitting processes will better enable communities to meet EPA’s 

National CSO Policy and stormwater permitting program requirements. 

iv 
Contents 

Notice ..............................................................................................................................................ii
 

Forward ..........................................................................................................................................iii
 

Abstract ..........................................................................................................................................iv
 

Tables............................................................................................................................................vii
 

Figures ...........................................................................................................................................ix
 

Acronyms and Abbreviations .......................................................................................................x
 

Acknowledgement......................................................................................................................xiii
 

Section 1 Introduction............................................................................................................ 1
 

Section 2 Conclusions ........................................................................................................... 7
 

Section 3 Recommendations ............................................................................................ 10
 

Section 4 Methods and Material ........................................................................................ 11
 

4.1 Literature Search.................................................................................... 11
 

4.2 Use of SWMM ........................................................................................ 19
 

4.3 Case Studies and Hypothetical Retrofit Examples ........................... 21
 

Section 5 Desktop Evaluation ............................................................................................ 22
 

5.1 Conversion of Primary Treatment Tanks............................................ 22
 

5.1.a. Augusta WWTF, Augusta, ME ................................................. 22
 

5.1.b. Hypothetical Retrofit of Primary Treatment Facilities 

Using the ACTIFLO System...................................................... 37
 

5.2	 Retrofitting Existing WWF Storage Tanks to Provide Enhanced 

Settling/Treatment and Post-Storm Solids Removal; 

Spring Creek AWPCP, New York, NY................................................. 53
 

5.2.a. Cross-Flow Plate Settlers.......................................................... 56
 

5.2.b. Chemical Addition to AWPCP .................................................. 63
 

5.2.c. Post-Storm Solids Removal...................................................... 69
 

5.3 Converting Dry Ponds for Enhanced Treatment .............................107
 

5.3.a. Sunnyvale Detention Basin, Santa Clara County, CA .......108
 

5.3.b. Monroe St. Detention Pond, Madison, WI............................121
 

5.3.c. Birmingham, AL Dry-Ponds ....................................................129
 

5.4 Retrofitting WWF Storage Tanks for DWF Augmentation.............139
 

5.4.a. Erie Boulevard Storage System, Syracuse, NY...................139
 

5.4.b. Spring Creek AWPCP, New York, NY...................................145
 

5.5 SSO Control Using Storage................................................................149
 

5.5.a. Hypothetical Overflow Retention Facility Retrofit to an 

Existing Sanitary Sewer System ............................................149
 

v 
Contents, Continued 

5.6	 Retrofitting for Industrial Wastewater Control in a
 

Combined Sewer System....................................................................156
 

5.6.a. Rockland WWTF, Rockland, ME ...........................................156
 

5.7	 Bringing Outdated/Abandoned POTWs Back Online as
 

Wet-Weather Flow Treatment Facilities............................................174
 

5.7.a. Auburn WWTF, Auburn, NY....................................................174
 

Section 6 References ..................................................................................................185
 

Appendices
 

Appendix A: Bibliographic Databases..........................................................191
 

Appendix B: Selected Tables with English Equivalent Units....................195
 

vi 
Tables 

2-1. Summary of Retrofitting Desktop Analysis Projects ............................................. 7
 

5.1.a-1 Phase 1 CSO Abatement Program Wet-Weather Treatment Design
 

Criteria........................................................................................................................ 30
 

5.1.a-2 Estimated O&M Costs, CSO Abatement Program – All Phases ...................... 35
 

5.1.a-3 Phase 1 CSO Abatement Program Project Costs............................................... 36
 

5.1.b-1 ACTIFLO Design Summary.................................................................................... 37
 

5.1.b-2 Comparison of ACTIFLO and Existing Primary Treatment Systems ............... 39
 

5.1.b-3 Average Annual Overflow Volume From the Pre- and Post-Retrofit
 

Facilities (1962-1991) .............................................................................................. 44
 

5.1.b-4 Estimated O&M Costs ............................................................................................. 46
 

5.1.b-5 Estimated ACTIFLO Retrofit Capital Costs .......................................................... 47
 

5.1.b-6 Major-Operating Parameters of ACTIFLO Pilot Plant at Nominal Flow Rate
 

of 150 M3/H................................................................................................................ 47
 

5.1.b-7 Removal Efficiency From 20 Daily Samples in Primary Treatment
 

in Mexico City, Mexico............................................................................................. 48
 

5.1.b-8 Effect of Coagulant Dosage on Removal Efficiencies in Galveston, Texas.... 50
 

5.1.b-9 Effect of Overflow Rate on Removal Efficiencies, Cincinnati, Ohio ................. 51
 

5.2-1 Existing Dimensions of the AWPCP and Design Flow Rate ............................. 55
 

5.2.a-1 Design Parameters for the Cross-Flow Plate Separator.................................... 56
 

5.2.a-2 Cross-Flow Plate Settler Design Data................................................................... 63
 

5.2.b-1 Coagulant Dosages and Performance Data for Six Treatment Plants
 

That Employ Chemical Coagulation for Advanced Primary Performance 

In Sedimentation Tanks........................................................................................... 66
 

5.2.b-2 Estimated Capital Costs Chemical Addition Retrofit........................................... 67
 

5.2.b-3 Estimated Chemical Addition O&M Costs ............................................................ 68
 

5.2.c-1 Work Plan Prototype Test Matrix ........................................................................... 76
 

5.2.c-2 Summary of Prototype Testing Operation............................................................ 82
 

5.2.c-3 Operating Data for Existing Traveling Bridge ...................................................... 89
 

5.2.c-4 Tipping Bucket Cleaning System Test Results at Representative
 

Solids Conditions ...................................................................................................... 90
 

5.2.c-5 Spray Nozzle Cleaning System Test Results ...................................................... 93
 

5.2.c-6 Original Orifice Header Configuration Test Results............................................ 95
 

5.2.c-7 Modified Orifice Header Configuration Test Results at Representative
 

Solids Concentrations .............................................................................................. 96
 

5.2.c-8 Projected Cleaning Requirements for a Single Basin Cleaning Event ..........102
 

5.2.c-9 Cost Comparison of Full Scale Tipping Bucket Versus Orifice
 

Header Cleaning Systems ....................................................................................104
 

5.3.a-1 Comparison of Median Metal Concentrations at Inlet to Retrofitted Basin
 

to Other Santa Clara Valley Stormwater Monitoring Station Data..................115
 

5.3.a-2 Inlet and Outlet Observed Concentrations and Pollutant Removals ..............116
 

5.3.a-3 Sediment Observations (mg/kg)...........................................................................117
 

vii 
Tables, Continued 

5.3.a-4 Comparison of Average Sediment Concentrations from Detention 

Basins and Swales (mg/kg) ..................................................................................118
 

5.3.a-5 Estimated Mean Annual Load Reduction and Cost Effectiveness .................119
 

5.3.a-6 Estimated Annualized Costs for Capital Expenditures and Operation...........120
 

5.3.b-1 Summary of Observed Influent and Effluent Pollutant Concentrations
 

at Monroe St. Pond ................................................................................................124
 

5.3.b-2 Summary Table of Pollutant Control...................................................................127
 

5.3.b-3 Average Suspended Solids Removal for Variable n Values for the
 

Monroe St. Ponds...................................................................................................128
 

5.3.c-1 Outlet Device Descriptions....................................................................................133
 

5.3.c-2 Subcatchment Summaries for Design Storms...................................................133
 

5.3.c-3 Pond Results of HydroCAD Simulations.............................................................134
 

5.3.c-4 DETPOND Summary for Design Storms............................................................135
 

5.3.c-5 Water Quality Output Summary for 1976 Rain File...........................................135
 

5.3.c-6 Water Quality Output Summary for 1952-1989 Rain File ................................138
 

5.4.a-1 Average Annual Capture Volume From the EBSS Under Pre- and
 

Post-Retrofit Conditions (1962-19991) ...............................................................143
 

5.4.a-2 Estimated O&M Costs ...........................................................................................144
 

5.4.a-3 Estimated EBSS Retrofit Capital Costs ..............................................................144
 

5.4.b-1 Wet-Weather Discharge Volume From the Spring Creek AWPCP
 

Under Pre- and Post-Retrofit Conditions for 1985 ............................................146
 

5.4.b-2 Estimated O&M Costs ...........................................................................................147
 

5.4.b-3 Estimated Spring Creek DWF Retrofit Capital Costs .......................................148
 

5.5.a-1 Average Annual Overflow Volume from the Pre- and Post-Retrofit
 

Facilities (1971-1985) ............................................................................................151
 

5.5.a-2 ORF O&M Costs.....................................................................................................153
 

5.5.a-3 Estimated ORF Construction Costs.....................................................................154
 

5.6.a-1 SWMM Modeling Summary – Annual Overflow Volumes................................164
 

5.6.a-2 SWMM Modeling Summary – Projected Peak Flows .......................................164
 

5.6.a-3 SWMM Modeling Summary – Projected Peak Flows .......................................165
 

5.6.a-4 Wet-Weather Treatment Design Criteria ............................................................167
 

5.6.a-5 Estimated O&M Costs Phase II CSO Improvements .......................................172
 

5.6.a-6 Project Costs – Phase II CSO Improvements....................................................173
 

5.7.a-1 Basis of Design of the Overflow Retention Facilities........................................178
 

5.7.a-2 Average Annual Overflow Volume From the Pre- and Post-Retrofit
 

Facilities (1971-1991) ............................................................................................181
 

5.7.a-3 ORF O&M Costs.....................................................................................................183
 

5.7.a-4 ORF Chlorination/Dechlorination Capital Costs ................................................183
 

viii 
Figures 

5.1.a-1 Existing Site Layout Prior to Retrofit...................................................................... 24
 

5.1.a-2 Process Flow Schematic Prior to Retrofit............................................................. 25
 

5.1.a-3 Site Layout Following Retrofit................................................................................. 28
 

5.1.a-4 Process Flow Schematic Following Retrofit......................................................... 29
 

5.1.b-1 Layout of Hypothetical WWTF Prior to Construction of ACTIFLO System..... 38
 

5.1.b-2 Layout of Proposed ACTIFLO System in Hypothetical Setting......................... 40
 

5.1.b-3 ACTIFLO Process Flow Diagram .......................................................................... 42
 

5.1.b-4 Helminth Eggs Removal in Primary Wastewater Treatment in Mexico
 

City, Mexico............................................................................................................... 49
 

5.1.b-5 SS Removal in Primary Wastewater Treatment in Mexico City, Mexico......... 49
 

5.1.b-6 Start-Up Efficiency of ACTIFLO Process.............................................................. 50
 

5.1.b-7 Twenty-Four Hour Continuous Demonstration Run, Cincinnati, Ohio ............. 51
 

5.2-1 Spring Creek AWPCP Layout................................................................................. 54
 

5.2.a-1 Layout of Shell Co. Cross-Flow Plate Settler, Montreal Canada...................... 57
 

5.2.a-2 Horizontal Projections of Inclined Cross-Flow Lamella Plates.......................... 60
 

5.2.a-3 Plan View of Aquarius Cross-Flow Plate Settler ................................................. 61
 

5.2.a-4 Profile of Aquarius Cross-Flow Plate Settler........................................................ 62
 

5.2.c-1 Wash Water Piping System Schematic ................................................................ 72
 

5.2.c-2 Tipping Buckets –Enlarged Plan............................................................................ 74
 

5.2.c-3 Tipping Buckets – Section A................................................................................... 75
 

5.2.c-4 Modified Plywood Training Wall Sche matic ......................................................... 78
 

5.2.c-5 Flushing Gate – Section B ...................................................................................... 79
 

5.2.c-6 Spray Nozzle Header – Section ............................................................................ 81
 

5.2.c-7 Original Orifice Header Configuration - Section A .............................................. 85
 

5.2.c-8 Modified Orifice Header Configuration – Section A ............................................ 86
 

5.3.a-1 Sunnyvale Pump Station No. 2 Showing Retrofit Structures...........................111
 

5.3.c-1 Stage v. Surface Area Curve ................................................................................132
 

5.3.c-2 Pond Stage v. Particle Residue Control.............................................................136
 

5.3.c-3 Rain Depth v. Particle Residue Control ..............................................................137
 

5.3.c-4 Rain Intensity v. Particle Residue Control..........................................................137
 

5.4.a-1 EBSS Layout ...........................................................................................................140
 

5.6.a-1 Existing Site Layout Prior to Retrofit....................................................................158
 

5.6.a-2 Process Flow schematic Prior to Retrofit............................................................159
 

5.6.a-3 Site Layout Following Retrofit...............................................................................162
 

5.6.a-4 Process Flow Schematic Following Retrofit.......................................................163
 

5.7.a-1 Existing Wastewater Treatment Flow Diagram .................................................176
 

5.7.a-2 Conceptual Wastewater and Sludge Treatment Flow Diagram......................180
 

ix 
Acronyms and Abbreviations 

ac-ft Acre-feet
 

ASD Augusta Sanitary District
 

AWPCP Auxiliary Water Pollution Control Plant
 

BATS Burnet Avenue Trunk Sewer
 

BMP Best Management Practice
 

BOD Biochemical Oxygen Demand
 

CCS Combined Collection System
 

CDM Camp Dresser & McKee
 

CEAM Center for Exposure Assessment Modeling
 

cfs Cubic Feet per Second
 

cm Centimeter
 

CN Curve Number
 

COD Chemical Oxygen Demand
 

COV Coefficient of Variation
 

CPE Comprehensive Performance Evaluation
 

CSO Combined Sewer Overflow
 

CSS Combined Sewer System
 

cyd Cubic Yards
 

CWA Clean Water Act
 

D Day
 

DAF Dissolved Air Flotation
 

DEP Department of Environmental Protection
 

DWF Dry Weather Flow
 

EA Environmental Assessment
 

EBSS Erie Boulevard Storage System
 

EPA U.S. Environmental Protection Agency
 

EMC Event Mean Concentration
 

FDS Flow Distribution Structure
 

FeCl3 Ferric Chloride
 

fps Feet per Second
 

ft Feet
 

gal Gallon
 

gpd Gallons per day
 

gpm Gallons per minute
 

ha Hectare
 

HP Horse power
 

h Hour
 

IDF Intensity Distribution Frequency
 

In. Inch
 

I/I Infiltration and/or Inflow
 

JSRS James Street Relief Sewer
 

kg Kilogram
 

x 
Acronyms and Abbreviations, Continued 

km Kilometer
 

kW Kilowatt
 

l Liter
 

lbs Pounds
 

m Meter
 

mA Milliampere
 

mm Millimeter
 

mm Micrometer
 

mi Mile
 

Metro Municipality of Toronto Metropolitan
 

METRO Syracuse Metropolitan Wastewater Treatment Facility
 

MG Million gallons
 

MGD Million gallons per day
 

mg Milligram
 

mg/L Milligrams per Liter
 

mg Microgram
 

min Minute
 

MIS Main Interceptor Sewer
 

ml Milliliter
 

msl Mean Sea Level
 

mt Metric ton
 

MTP Main Treatment Plant
 

NASA National Aeronautics and Space Administration
 

NOAA National Oceanic and Atmospheric Administration
 

NPDES National Pollution Discharge Elimination System
 

NTIS National Technical Information Services
 

NURP Nationwide Urban Runoff Program
 

NWS National Weather Service
 

NYCDEP New York City Department of Environmental Protection
 

NYSDEC New York State Department of Environmental Conservation
 

O&M Operation and Maintenance
 

ORF Overflow Retention Facility
 

POTWs Publicly Owned Treatment Plants
 

ppm Parts per Million
 

PPS Particulate Pollutant Strength
 

PRF Peak Reduction Factors
 

psi Pounds Per Square Inch
 

RCP Reinforced Concrete Pipe
 

rpm Revolutions per Minute
 

s Seconds
 

SCADA Supervisory Control and Data Acquisition
 

scfm Standard Cubic Feet Per Minute
 

xi 
Acronyms and Abbreviations, Continued 

SCS Soil Conservation Service
 

SESD South Essex Sewerage District
 

SOR Surface Overflow Rate
 

STLC Soluble Threshold Limit Concentrations
 

SS Suspended Solids
 

SSES Sewer System Evaluation Survey
 

SSO Sanitary Sewer Overflow
 

SW Storm water
 

SWMM Stormwater Management Model
 

SYNOP Synoptic Rainfall Analysis Program
 

TB Tipping Buckets
 

TKN Total Kjeldahl Nitrogen
 

TOC Total Organic Carbon
 

TTLC Total Threshold Limit Concentration
 

USGS U.S. Geological Survey
 

WET Whole Effluent Toxicity
 

WHO World Heath Organization
 

WPCP Water Pollution Control Plant
 

WQO Water Quality Objective
 

WWF Wet Weather Flow
 

WWTF Wastewater Treatment Facility
 

xii 
Acknowledgements 

Moffa & Associates gratefully acknowledges the support of the project by the National 

Risk Management Research Laboratory, Office of Research and Development, U.S. 

Environmental Protection Agency, under the direction of Mr. Richard Field, Wet-Weather 

Flow Research Program leader. 

The cooperation of Thomas P. O’Connor, Project Officer, Urban Watershed Management 

Branch, Water Supply and Water Resources Division, National Risk Management 

Research Laboratory (NRMRL), Office of Research and Development (ORD), U.S. 

Environmental Protection Agency (EPA), is acknowledged as are Dr. David Fischer and 

Dr. Ronald Rovanesk, ORISE post graduates with EPA’s UWMB, in Edison, New Jersey 

and Dan Murray, Technology Transfer Division, NRMRL, ORD, EPA, Cincinnati, Ohio for 

providing reviews for the report. 

In addition, special thanks are given to the following consultants for their contributions to 

this report: 

Steve D. Freedman, P.E., Earth Tech Inc. 

Sorin L. Goldstein, P.E., Camp Dresser & McKee, Inc., Woodbury, NY 11797 

Dwight A. MacArthur, P.E., O’Brien & Gere Engineers, Inc., Syracuse, NY 13221 

Robert Pitt, Environmental Engineer, Consultant 

Robert B. Stallings, P.E., Earth Tech Inc. 

The assistance of Gregory J. Hotaling with preparation of graphical materials presented is 

hereby recognized, as is the assistance of Renee Davis for assistance with word 

processing. 

This report was prepared by Moffa & Associates Consulting Engineers, by Howard M. 

Goebel, Daniel P. Davis, and John J. LaGorga under the direction of Peter E. Moffa, 

Principal. 

xiii 
Section 1 

Introduction 

Background 

Wet-weather flow (WWF) is currently one of the leading causes of water-quality 

impairment in the United States and improvement of controls is one of the two priority 

water focus-areas cited by the U.S. Environmental Protection Agency’s (EPA) Office of 

Water in its National Agenda for the Future. Wet-weather pollution problems are 

primarily manifested in three different ways: separate sewer overflows (SSO), combined 

sewer overflows (CSO), and stormwater (SW) discharges. SSO are the result of the 

unplanned relief of a sewer system intended only for sanitary sewage but also carrying 

infiltration and inflow (I/I), illegal stormwater connections, and stormwater. CSO are the 

result of the “designed” relief of a sewer system that intentionally carries stormwater and 

sanitary sewage. Stormwater runoff discharges include flows that have been collected 

in a separate storm sewer system or a surface drainage system. 

WWF pollution is extensive throughout the country. There are approximately 15,000 

CSO discharge points within 1,100 municipalities that have combined sewers. SSO 

occur in more than 1,000 municipalities. SW discharges occur in over 6,500 

municipalities with populations greater than 50,000 and at 1.2 million industrial, 

commercial, institutional, and retail sources. 

It was not until data from several comprehensive municipal CSO and SSO studies were 

collected that the types of pollutants discharged by overflow systems and their relative 

impacts on receiving waters were initially understood. The primary parameters of 

importance are visible floatables and gross solids, bacteria and other pathogens, 

suspended solids (SS), biochemical oxygen demand (BOD), nutrients, and toxic 

organics and metals. 

The projected costs for national CSO, SSO, and SW abatement are in the tens of 

billions of dollars. Consequently, inexpensive abatement alternatives are required to 

alleviate the high costs associated with conventional treatment. High-rate treatment 

technologies can increase sewage treatment system capacity to allow treatment of a 

significant portion of the WWF in sanitary and combined sewers. The primary benefits 

of high-rate treatment technologies are their ability to handle higher flow rates and 

compactness or small footprint compared to conventional treatment processes. Highrate treatment technologies require less tankage and space and in urbanized areas are 

often more cost effective than conventional treatment processes since storm flows are 

significantly greater than dry-weather flows. 

Wet-weather discharges are reflections of the watersheds or sewersheds from which 

they are generated and the climatic patterns those areas experience. The impacts of 

wet-weather discharges are determined by a variety of complex relationships between 

the temporal and spatial distributions of discharges. These include the types and 

magnitudes of pollutant loads being transported, the ecological and hydraulic nature of 

the receiving water, as well as the designated beneficial uses, and the desires and 

1
 
expectations of stakeholders. The result is a wide-scale problem that requires very sitespecific solutions. 

Solutions to WWF problems require investigation of a wide variety of management 

options and the relationships between management costs and water quality benefits. 

WWF must be controlled by storage-treatment systems since WWF are intermittent and 

highly variable in both pollutant concentrations and flow rates. An optimized system 

maximizes the use of the existing infrastructure prior to construction of new 

treatment/storage devices. 

Retrofitting existing sewerage systems to handle increased WWF that result in SSO, 

CSO, and SW discharges may be a cost-effective alternative to the construction of new 

treatment systems. Retrofitting existing sewerage systems and treatment plants to 

convey and treat additional WWF can be accomplished by: 

1. Increasing the hydraulic loadings at the control facilities (CSO and SSO); 

2.	 Increasing the amount of storage throughout the conveyance system (All WWF); 

and 

3.	 Providing the additional function of pollution removal by settling at flood-controldetention basins (SW). 

Summary of Analyses 

This report demonstrates real and hypothetical case studies. Retrofits of available 

technologies for conventional sewer systems are presented in the form of desktop 

analyses. The desktop analyses were conducted to determine the technical feasibility 

and cost effectiveness of retrofitting existing sewage and WWF systems in lieu of 

constructing new sewerage and conveyance systems. Site-specific analyses were 

completed for WWF storage-treatment systems and their associated cost due to the 

variability of wet-weather volume and treatability (quantity and quality). 

The desktop analyses including conceptual layouts and designs of systems to increase 

treatment and storage of WWF through retrofits were performed for the following 

scenarios: 

1.	 Converting or retrofitting primary settling tanks at the publicly owned treatment 

works (POTW) to alternative sedimentation or sedimentation enhancement 

methods or methods of treatment that provide equivalent or higher levels of 

treatment (SSO and CSO); 

2.	 Retrofitting existing CSO storage tanks to provide enhanced settling/treatment 

and post-storm solids removal; 

3.	 Converting existing “dry-ponds” and/or other flood control devices to SW 

detention ponds for separate SW systems to allow treatment through 

sedimentation; 

4.	 Retrofitting existing WWF storage tanks and ponds to serve a second function as 

DWF equalization tanks; 

2
 
5. Retrofitting for industrial wastewater control in a combined sewer system; 

6. SSO control using storage; and 

7.	 Bringing outdated and/or abandoned POTWs back online as satellite WWF 

treatment facilities (SSO and CSO). 

The analyses were performed over a variety of orders of magnitude of flow and/or 

volume for each scenario. The desktop analyses of the retrofitted systems addressed 

the following design considerations, where appropriate: 

1. Operations, maintenance and level of automation 

2. Sludge handling requirements and disposal 

3. Continuous operation versus storm flow event operations 

4. Chemical storage and feed rate 

5. Bypass requirements 

Treatment values, pollutant and hydraulic loadings, and seasonal and diurnal variation 

are also provided, where appropriate. 

Simplified simulations were completed using historical rainfall data for an one-year 

period using the EPA Stormwater Management Model (SWMM) with hourly time steps 

or an appropriate alternative. The yearly rainfall data is provided including seasonal 

variation, average antecedent duration, and storm intensity. 

When feasible, the cost effectiveness of retrofitting versus construction of additional 

conventional treatment works and/or storage basins was completed as part of each 

desktop analysis. The effectiveness of each system, guidance on design and operation, 

cost breakdowns (new construction and modification of existing structure, system 

installation, and operation and maintenance [O&M]), and case studies are provided. 

Regulatory Background 

Since the 1970s, the U.S. Environmental Protection Agency’s Construction Grant 

Program financed the development and construction of hundreds of wastewater 

treatment facilities (WWTF) around the nation. During this time the main focus was 

regional WWTF designed to treat dry-weather sewage flows. Little or no money was 

invested to abate wet-weather pollution sources. Since then, many of the POTWs have 

exceeded their design flows, discharge standards have become more stringent, and 

federal regulations and policies have begun to address WWF pollution. All of these 

factors contribute to a need for more and better WWF treatment facilities. Today, 

federal funding for new environmental facilities is scarce and taxpayers carry the 

burden. Based on this climate, it becomes much more critical that any compliance 

needs should first consider optimizing existing infrastructure through retrofitting. 

3
 
This report is intended to help municipalities who may be impacted by current policies 

and regulations (i.e., National Pollutant Discharge Elimination System (NPDES) Phase I 

permitting and the CSO Control Policy) and the newer regulations and policies that 

address WWF (i.e., the upcoming NPDES (Phase II) permitting regulations and SSO 

regulations). Also, this report can assist municipalities and others (e.g., industry and 

commercial property owners) who are subject to local regulations. 

Stormwater Regulations and Permits 

Congress added Section 402(p) to the Clean Water Act in 1987 to require the 

implementation of a comprehensive approach for addressing stormwater discharges. 

Section 402(p)(4) required EPA to develop permit application regulations under the 

NPDES, submission of NPDES permit applications, issuance of NPDES permits, and 

compliance with NPDES permit conditions. Section 402(p)(6) requires EPA to designate 

stormwater discharges to be regulated (within the statutory definitions provided in 

section 402(p)(2)) and establish a comprehensive regulatory program, which may 

include performance standards, guidelines, guidance, and management practices and 

treatment requirements (EPA, 1999). 

The first phase of NPDES stormwater permit application regulations (“Phase I”) was 

promulgated on November 16, 1990 (EPA, 1990). The provisions addressing municipal 

separate-storm-sewer systems (MS4s) cover systems serving a population of 100,000 

or more. This includes 173 cities, 47 counties and additional systems designated by 

EPA or states. A total of 260 permits, covering approximately 880 operators (local 

governments, state highway departments, etc.) were identified as subject to Phase I 

permit application requirements. The CWA requires that MS4 permits effectively 

prohibit non-stormwater discharges into the storm sewers as well as reduce the 

discharge of pollutants to the maximum extent practicable (including management 

practices, control techniques and system, design and engineering methods, and other 

provisions appropriate for the control of such pollutants). 

Phase I MS4 permittees were required to submit an application that included sourceidentification information, precipitation data, existing data on the volume and quality of 

stormwater discharges, a list of receiving water bodies and existing information on 

impacts on receiving waters, a field screening analysis for illicit connections and illegal 

dumping, and other information. The MS4 permittees were to gather and provide 

additional information including: 

• discharge characterization data based on quantitative data from 5 to 10 representative 

locations in approved sampling plans; estimates of the annual pollutant load and eventmean concentration of system discharges for selected conventional pollutants and 

heavy metals; a proposed schedule to provide estimates of seasonal pollutant loads; 

and the mean concentration for certain detected constituents in a representative storm 

event; and 

4
 
• a proposed management program including descriptions of: structural and source 

control measures that are to be implemented to reduce pollutants in runoff from 

commercial and residential areas; a program to detect and remove illicit discharges; and 

a program to control pollutants in construction site runoff. 

The NPDES stormwater regulations for the second phase of stormwater discharge
 

control (“Phase II”) was proposed on January 9, 1998 (US EPA, 1998c). EPA was
 

required to promulgate the Phase II rule in 1999 under a separate consent decree. The
 

proposal designated two classes of facilities to be automatically covered on a
 

nationwide basis under the NPDES program:
 

(1) small municipal separate storm sewer systems located in urbanized areas (about
 

3,500 municipalities would be included in the program); and 

(2) construction activities (pollutants include sediments and erosion from these sites) 

that disturb equal to or greater than one and less than five acres of land (about 

110,000 sites per year will be included in the program). 

Those facilities designated above would need to apply for NPDES stormwater permits 

by 2002. EPA is anticipating that most permittees would be covered under general 

permits. 

Municipalities and others may be able to use retrofitting to comply with the new Phase II 

rules by converting existing dry ponds (or a portion thereof), used typically for flood and 

drainage control, to wet ponds. This is consistent the Phase I management program to 

use structural and source control measures to reduce pollutants in runoff from 

commercial and residential areas. Wet ponds, because of the permanent pool and 

longer detention times are thought to remove more pollutants than dry ponds (e.g., SS 

and BOD and COD removals) and reduce or eliminate discharges. 

Some facilities that EPA is proposing to cover under the new Phase II rule are already 

subject to state and/or local stormwater management requirements. The Rockland, 

Maine case study demonstrates how retrofitting was incorporated into the Maine's CSO 

Abatement Program permitting requirements. Among other planned retrofits a new 

force main was commissioned to separate an industrial loading from the municipal 

sewer and the wastewater treatment facility is being upgraded to handle additional 

WWF. 

CSO Control Policy 

The need for high-rate treatment systems comes from the National CSO Policy that 

requires treatment of additional WWF in existing POTW during high flow periods caused 

by rainstorms. The recent EPA National CSO Control Policy (59 Federal Register 

18688) resulted in the publication of several guidance documents. These two guidance 

documents are “Combined Sewer Overflow – Guidance for Nine Minimum Controls” 

(EPA 832-B-95-003) and “Combined Sewer Overflow – Guidance for Long Term Control 

Plan” (EPA 832-B-95-002). Required CSO abatement technologies that are addressed 

in this report are: 

5
 
¤ Maximization of flow to the POTW for treatment; 

¤	 Evaluation of alternatives that enable the permittee, in consultation with the NPDES 

permitting authority, Water Quality Standard (WQS) authority, and the public, to 

select CSO controls that will meet CWA requirements; 

¤	 Cost/performance considerations to demonstrate the relationships among a 

comprehensive set of reasonable control alternatives; and 

¤ Maximization of WWF treatment at existing POTW treatment plants 

Consensus Recommendation of SSO Federal Advisory Subcommittee 

The EPA/State SSO work group distributed a set of draft NPDES permit regulations and 

related guidance for municipal sanitary sewer collection systems and SSO to a small 

municipal outreach group, the SSO Federal Advisory Subcommittee and other members 

of the public. The SSO Federal Advisory Subcommittee made consensus 

recommendations to EPA on: 

• capacity, management, operation and maintenance (CMOM) program requirements; 

• a permit prohibition for SSO; 

• permit record keeping, reporting and public notification requirements; and 

• peak excess flow treatment facilities. 

The SSO subcommittee recommended to EPA the substance of the CMOM, Prohibition, 

Record Keeping, Reporting and Public Notification, Remote Treatment Facilities 

documents, and Satellite Collection Systems and watershed management agreed 

principals on October 20, 1999. In addition, the Subcommittee recommended principles 

for satellite collection systems and watershed management. 

The protocols for the Federal Advisory Subcommittee provide that EPA is committed to 

reflect the consensus reached by the SSO Subcommittee in an upcoming notice of 

proposed rulemaking. While the Advisory Subcommittee made recommendation 

addressing core issues, some outstanding issues remain. The consensus 

recommendation of the SSO Federal Advisory Subcommittee clears the way for the 

next step in the development of a proposed rule. 

Retrofitting an existing treatment plant is often a more cost-effective way to control 

additional flow than building new facilities. With the upcoming SSO regulations, 

investigating the economics of retrofitting existing systems to treat more flow, may 

become the first step in a compliance exercise. 

6
 
Section 2 

Conclusions 

Desktop analyses were performed for 13 separate retrofit examples. The analyses 

demonstrated that in certain circumstances retrofitting existing WWF facilities is 

technically feasible and may be more cost effective than construction of new 

conventional control and treatment facilities. A summary of each retrofit is provided in 

Table 2-1 and is discussed below. 

Table 2-1. Summary of Retrofitting Desktop Analysis Projects 

Project Retrofit Treatment Cost 

Benefit Benefit 

Augusta WWTF Headworks Upgrade - Enhanced 

Primary Treatment 

Yes Yes 

Hypothetical WWTF ACTIFLO System -Enhanced Primary 

Treatment 

Yes No 

Spring Creek AWPCP Plate Settlers for Enhanced 

Treatment in WWF Storage Facilities 

No No 

Spring Creek AWPCP Chemical Addition for Enhanced Marginal Marginal 

Treatment in WWF Storage Facilities 

Spring Creek AWPCP Post-Storm Solids Removal for WWF Yes Not 

Storage Facilities Determined 

Sunnyvale Detention Basin Retrofit for Enhanced Treatment Yes Not 

Determined 

Monroe St. Detention Pond Retrofit for Enhanced Treatment Yes Not 

Determined 

Birmingham Dry-Ponds Retrofit for Enhanced Treatment Yes Not 

Determined 

Erie Boulevard Storage System DWF Augmentation Yes Marginal 

Spring Creek AWPCP DWF Augmentation Yes Marginal 

Hypothetical Sanitary Sewer 

System 

SSO Control Using Storage Yes Yes 

Rockland WWTF Retrofitting for Industrial Wastewater Yes Yes 

Control in a CSS 

The headworks of the Augusta, ME WWTF were retrofitted to maximize flow to the 

existing primary and secondary tanks and to minimize wet-weather bypass. This was a 

very cost-effective alternative by maximizing existing plant capacity. This retrofit 

allowed for minimizing the construction of new high-rate treatment facilities. However, 

there are only limited experiences of such retrofitting documented in the literature. 

Enhanced sedimentation methods employ flocculation with lamellar settling to speed up 

the clarification process. One such system, the ACTIFLO, was hypothetically retrofitted 

to a primary treatment facility. This technology provides enhanced SS removal with 

short detention times and quick start-up as compared to conventional primary 

clarification. One of the benefits of the ACTIFLO system is the extremely small footprint 

7
 
of the process. The downside is the high cost of chemicals and increased O&M 

requirements. 

Retrofitting existing WWF storage tanks for flow equalization and enhanced treatment 

was not cost effective. The Spring Creek Auxiliary Wastewater Pollution Control Plant 

(AWPCP) in New York, NY was used to evaluate this retrofit option. This is largely due 

to the high variability of WWF and limitations of the technologies evaluated. Retrofitting 

existing WWF storage tanks with cross-flow plate settlers was found to be technically 

limiting due to the hydraulic loading rates of cross-flow plate settlers. This technology 

may be more appropriate for primary clarifiers where hydraulic load rates are more 

consistent than in WWF storage tanks. 

Retrofitting existing WWF storage tanks with chemical addition to provide chemically 

enhanced sedimentation of WWF is a feasible option. This technology could result in 

60—90% SS removal, 40—70% BOD5 removal, and 30—60% chemical oxygen 

demand (COD) removal. This technology does not affect the operation of the storage 

tanks to serve their primary role of equalization. 

The AWPCP was also used to evaluate several retrofitting options for post-storm solids 

removal from WWF storage tanks including tipping buckets, flushing gates, and orifice 

headers. The results of pilot testing and design considerations including capital and 

O&M costs are provided for each technology. 

Retrofitting existing dry-ponds to provide treatment through sedimentation was found to 

be a relatively simple and viable treatment option. Three dry ponds were evaluated 

from different regions of the country: 1) Santa Clara, CA, 2) Madison, WI, and 3) 

Birmingham, AL. The retrofits significantly reduced effluent flows and downstream 

channel erosion and also provided a water quality benefit. This was done largely by 

developing features to better control influent and effluent flows and to reduce shortcircuiting within the ponds. 

The Erie Boulevard Storage System (EBSS) in Syracuse, NY and the AWPCP were 

evaluated for the purpose of serving a dual role as wet-weather and dry-weather 

equalization facilities. This analysis showed that this could be done without drastically 

reducing the ability of the storage tanks to capture WWF. The primary benefit of 

equalization provided by the DWF storage is the dampening of peak flow rates that 

would otherwise result at the treatment plant, allowing the primary and secondary 

treatment facilities to operate at a more uniform rate thereby provide more consistent 

treatment efficiency throughout a typical day. However, this retrofit is only economically 

viable for a treatment plant that is operating at or near design capacity where increased 

O&M costs can be balanced against cost to increase capacity at the plant. 

The retrofit of an unused aeration tank at the Rockland WWTF in Rockland, ME and the 

construction of a separate force main from an industrial facility to the WWTF provided a 

cost-effective alternative for the elimination of high-strength wastes from CSO 

discharges. This retrofit conveyed high-strength waste from the industrial facility to the 

8
 
dry-weather headworks thereby eliminating the potential for CSO discharges of 

industrial wastewater and consequent impact on the receiving water. 

Retrofitting SSOs to use storage from overflow retention facilities provided a costeffective alternative to expansion of the sanitary collection system to convey the volume 

of overflow to the treatment plant for primary and secondary treatment. However, this 

alternative requires the availability of a sufficient amount of land to construct the storage 

facility. 

Retrofitting abandoned POTWs for use as WWF treatment facilities was shown to be an 

effective treatment alternative. The Auburn WWTF in Auburn, NY was retrofitted using 

the abandoned primary treatment tanks as WWF storage tanks to reduce the impact 

associated with WWF in excess of the capacity of the existing treatment plant. This 

retrofit was an extremely cost-effective alternative to other abatement options due to the 

availability and condition of the existing facilities. The abandoned primary tanks were in 

good condition and required only minor modifications. This retrofit is a cost-effective 

alternative in cases where abandoned primary tanks are available in good condition for 

retrofitting. This type of retrofit may be a less cost-effective solution for cases where the 

existing tanks are in poor condition requiring a substantial cost to upgrade the tanks to 

serve a functional role. 

The feasibility and cost effectiveness of retrofitting is in large part controlled by sitespecific conditions and treatment requirements. The physical condition of existing 

facilities, age, and level of maintenance, size, layout, hydraulic factors, water quality and 

other site-specific factors all affect the suitability for retrofitting. As such, the opportunity 

for retrofitting existing WWF facilities must be evaluated on a case by case basis. 

9
 
Section 3 

Recommendations 

The findings of this study indicate that retrofitting existing control facilities can be a costeffective alternative to the construction of conventional treatment facilities. The 

appropriateness of retrofitting is however, a function of site-specific needs and 

conditions. Therefore, it is recommended that retrofitting existing control facilities be 

identified as an alternative for wet-weather abatement. Retrofitting should be evaluated 

early on in the decision making process for wet-weather abatement. 

The advanced sedimentation methods such as the ACTIFLO merit further study given 

the results for high-rate treatment applications in Europe and due to the limited 

experience in North America. This could include controlled, side-by-side pilot 

demonstrations to assess the benefits and cost effectiveness of alternative advanced 

sedimentation methods. 

10
 
Section 4
 

Methods and Materials
 

4.1 Literature Search 

Introduction 

Typically, urban infrastructure collects, conveys, and treats DWF. In the case of a CSO 

community, a component of WWF is collected and conveyed in the same infrastructure 

as the DWF, but points of relief are purposely provided before treatment because of 

hydraulic constraints of the conveyance system and the capacity of the dry-weather 

treatment facilities. In the case of an SSO community, the DWF (sanitary sewage) and 

the stormwater are collected and conveyed in different systems. However, these 

separated sanitary systems inadvertently receive I/I which result in overflows and 

consequent pollutant concerns. Regardless of how WWF are manifested, the essential 

infrastructure is made up of the collection, conveyance, and treatment systems. In each 

of these systems, retrofitting specific structures can provide improved control and 

treatment for WWF. Often surprisingly simple and inexpensive retrofits can provide 

expanded capacity, enhanced treatment, extended facility lifespan, and reduced 

operating costs. 

This literature review focuses on retrofits made to collection, conveyance, and treatment 

systems in order to better control and treat wet-weather pollution. A majority of the cited 

literature was found through queries in such databases as, 

• Environmental Sciences & Pollution Management Database 

• Compendex*Plus 

• Wilson Applied Science & Technology Abstracts 

• ProCite Database, 

• National Center for Environmental Publication and Information 

Other literature was found through requests for information made to consulting 

engineering firms. An emphasis was placed on national conference proceedings. The 

referenced databases are described in Appendix A. 

General Retrofitting Information 

Martin (1988) developed a method to help government watershed planners and 

managers to: (1) assess the pollutant potential of runoff from existing urban 

development, (2) screen for feasible control measures, and (3) develop and implement 

retrofit management strategies. This method strongly prompts the implementation of 

watershed retrofits as remedial actions for stormwater pollution. 

Walesh (1992) discusses the potential for retrofitting stormwater facilities. The paper 

summarizes the history of urban stormwater detention facilities and suggests that many 

of these existing facilities can be retrofitted to improve quantity control, include quality 

11
 
control, improve O&M, reduce safety hazards, enhance aesthetic attributes and have 

recreational features. Beale (1992) and England (1995) also discuss retrofitting 

stormwater facilities. Both authors address design considerations for stormwater 

retrofitting techniques. Such techniques include detention ponds, retention ponds, 

underground vaults, exfiltration pipes, and baffle boxes. Palhegyi, Driscoll, and 

Mangarella (1991) evaluated the feasibility of retrofitting stormwater control facilities to 

improve pollutant-removal performance. This paper presents the results of an analysis 

that was performed to predict the removal efficiency of suspended sediments in storm 

runoff under existing conditions and under several practical retrofit alternatives. A flood 

risk analysis was performed for each condition. It was concluded that stormwater 

quality could be improved by modifying operating strategies of existing facilities, but a 

flood risk analysis must be considered. 

Field and O’Connor (1997) discuss a strategy to optimize CSO control systems. The 

optimization strategy maximizes the use of the existing system and sizes the storage 

facility in accordance with the WWTF capacity to obtain the lowest-cost storage and 

treatment system. Roesner and Burgess (1992) discuss CSO rehabilitation strategies 

for urban areas. Roesner and Burgess address and cite examples for eliminating dryweather overflows, rehabilitating regulators, and optimizing existing interceptor capacity. 

Collection System Retrofits 

The practice of utilizing inlet control concepts to reduce the impacts of flooding from 

urban runoff and pollution through catchbasin adjustments is common. Most 

installations involve restriction of catchbasin outflow to cause surface inflow to be redirected by overland routes to more suitable discharge points or to underground 

storage. The majority of installations involve retrofitting existing systems to mitigate 

collection-system surcharging, reduce exfiltration and infiltration exchange within 

adjacent sewer systems, and to cause reduction of downstream CSOs (Pisano, 1990). 

The area known as the “Triangle” in Parma, Ohio, is a 12.1 ha (30 acre) low-relief area 

in a 117.4 ha (290 acre) drainage system. In the “Triangle” storm drains are generally 

inadequate and severe street flooding and basement flooding are common. Inlet 

controls such as overland flow berms, vortex throttles, and shallow drains to de-water 

street areas were implemented throughout the entire 117.4 ha (290 acre) drainage area. 

The project mitigated surface-water ponding and provided basement-flooding protection 

throughout the entire 117.4 ha (290 acre) area. Construction costs in 1984 were 

$875,000 (Pisano, 1982). 

Lake Quinsigamond, Worchester, Massachusetts, was becoming eutrophic from 

stormwater discharges. Studies indicated that the dominant nutrient source was 

contaminated DWF from the storm sewer. It was believed that this contamination 

resulted from the mixing of the sanitary sewer with the storm sewer due to the 

“over/under” sewer configuration. In this configuration the sanitary sewer was placed 

over the stormwater sewer in the same trench. A system wide pipe replacement or 

rehabilitation was rejected based on cost. Therefore, a number of system controls were 

12
 
installed; these were found to immediately eliminate much of the nutrient loading from 

the storm sewer base flow. A vortex flow throttle was installed in the storm sewer to 

divert the polluted base flow and the first flush to a nearby combined sewer. Inlet 

controls were then installed in the upper reaches of the combined system to increase its 

capacity to convey the polluted base flow and the first flush. Approximately 0.34 m3/s 

(12 cfs) of stormwater (base flow and first-flush) were diverted to the combined system. 

Total construction cost were estimated at $160,000 (Pisano, 1989). 

Conveyance System Retrofits 

The conveyance system is the infrastructure component designed to move water from 

the site of origin to an acceptable receiving water body. In the case of dry-weather 

sewage flow, the conveyance system carries wastewater to a treatment plant. In the 

case of WWF, this water may be discharged into a water body without treatment. In 

many stormwater instances, detention ponds are an integral part of the conveyance 

system. In the past, these detention ponds were used primarily for the purpose of urban 

flood control. In some instances, ponds are retrofitted for the capture and treatment of 

stormwater, thereby improving the quality of the receiving water body. 

Flow Equalization 

Fairport Harbor, Connecticut, developed a unique solution to eliminate CSOs by 

converting an abandoned industrial fuel oil tank into a cost-effective wet-weather 

retention tank. The fuel oil tank was utilized for flow equalization after the original plan 

to construct a new tank and pump station near the outfall met local resistance. The 

conversion of the fuel oil tank has removed an environmental liability, saved the district 

money, and reduced CSOs. An initial examination of the 36.6 m (120 ft) diameter fuel 

oil tank indicated it had sufficient capacity 12,112 m3

 (3.2 MG) for the intended capture 

of the five-year intensity storm. However, of primary concern were structural integrity 

and the presence of industrial sludge and asbestos. Rehabilitation of the tank began 

with the removal of the crude oil sludge, interior piping, asbestos-covered exterior 

piping, and lead-based paint. Welds that did not meet American Welding Society 

standards were replaced. Utilization of the fuel oil tank also required building a pump 

station and force main designed for a peak flow rate of 19,870 m3/d (5.25 million gallons 

per day or MGD). The project total cost was $2.6 million (Shrout, 1994). 

The Rohnert Park Demonstration Project included the design, construction, operation, 

testing and evaluation of a surge facility designed to provide flow equalization and some 

degree of treatment to all WWF and to provide rate control of wet-weather and dryweather wastewater flows to interceptor sewers. The principal feature of the surge 

facility was a 2840 m3 (0.75 million gallon or MG) sedimentation-equalization basin. It 

was tested under field conditions with influent flows varying from 2,270—20,800 m3/d 

(0.6—5.5 MGD). The ability of the surge facility to provide adequate hydraulic control to 

diurnal flow variation was documented. Under WWF conditions, the surge facility was 

able to remove approximately 45% of the influent SS. BOD removal under WWF 

13
 
conditions was not significant. The total construction cost for the demonstration facility 

was $384,000 (Welbon, 1974). 

Detention Basins 

Santa Clara Valley, California is protected from flooding by a series of pump stations 

and levees. Based on a preliminary evaluation of the feasibility of retrofitting detention 

basins, a pilot study was implemented to retrofit a pump station and to conduct testing 

to measure water quality benefits and costs. In the pilot study, structural and 

operational retrofits were implemented at a pump station consisting of four primary 

pumps rated at 1.1 m3/s (39 cfs) and one auxiliary pump rated at 0.25 m3

/s (9 cfs). The 

detention basin area is 1.8 ha (4.4 ac.) with a 37,005 m3

 (30 ac-ft) capacity and receives 

water from a 187.4 ha (463 acre) urbanized watershed. The structural retrofits included 

installing a weir at the outlet of the detention pond, filling a channel that connected the 

inlet and the outlet of the detention pond, and blocking off a drainage pipe that ran 

under the channel. The purpose of these structural retrofits was to reduce shortcircuiting of the pond and provide better distribution of the flow into the outlet. The 

operational retrofit consisted of modifying the pump schedule to create a two-foot 

permanent pool in the pond. These retrofits provided greater retention time in the 

detention pond and therefore enhanced gravitation settling of SS. Post-retrofit water 

quality surveys showed removals of 29% for chromium, 42% for total copper, 53% for 

lead, 51% for total nickel, 44% for total zinc, and 50% for SS. The metals removals 

correlated well with the SS removal. The amortized cost over 20 years for retrofitting, 

operation, and maintenance was $8,200/year (Woodward-Clyde, 1996). 

Dye-tracing experiments performed in a stormwater pond in Kingston Township, Ontario 

revealed short retention times, mainly due to the small size of the pond. Mathematical 

modeling of this pond indicated irregular circulation patterns, resulting in short-circuiting 

and dead zones. A study was conducted to retrofit the pond for the purpose of 

improving pollutant-removal characteristics. A series of three baffles were installed 

along the length of the pond, roughly perpendicular to the direction of flow, and 

extending an average of two-thirds of the width across the pond. The installation of the 

baffles increased the length-to-width ratio of the flow path in the pond from 1.5:1 to 4.5:1 

and corrected the flow regime problems. In addition, an increase in the hydraulic 

efficiency of the pond (defined as the ratio of measured to volumetric retention times) 

from 0.65—0.86 was observed. An increase in pollutant removal through sedimentation 

processes was inferred from a comparison of retention time measured before and after 

baffle installation (Matthews et al., 1997). 

Florida’s Indian River Lagoon suffers from drainage problems such as increased volume 

of freshwater runoff to the estuarine receiving water and deposition of organic 

sediments, reduced water clarity because of increased discharge of SS, and 

eutrophication caused by nutrient loadings. A project was initiated to create a 

watershed control system for the Indian River Lagoon, develop management strategies 

to relieve stresses resulting from runoff to the lagoon, and address the feasibility of 

watershed retrofitting to reduce pollutant loads. The partially developed, 768.9 ha 

14
 
(1,900 acre), industrial area of NASA’s Kennedy Space Center was selected as a 

representative drainage catchment. After screening the retrofitting alternatives, it was 

determined that constructing a weir in the main drainage channel, with discharge of 

diverted flow to a wetland area, would provide the best system performance. The cause 

and effect relationships between the catchment hydrology, channel hydraulics, and 

pollutant loads were documented using a calibrated SWMM model. Model results 

concluded that the retrofit could satisfactorily achieve flood control, increase SS 

removal, and maximize water depth, but could have difficulty meeting the groundwater 

discharge and water level-fluctuation criteria. Without periodic drawdown from the 

wetland, water levels in the system would be nearly static due to the nearly flat relief. 

Fixed cost of the proposed retrofit would be $28,000. The annual O&M costs were 

estimated to be $17,100 (Bennett and Heaney, 1989). 

Chemical Enhanced Treatment 

As evidenced by the above-referenced literature, many structural retrofits have been 

implemented to improve the treatment characteristics of wet-weather detention basins. 

In addition to structural retrofits, some municipalities have experimented with chemically 

enhanced wet-weather detention basins to improve the treatment characteristics of 

these basins during wet-weather events. This low-structural retrofit requires small 

amounts of chemicals and polymer to promote the coagulation and flocculation of 

smaller, lighter particles into larger, heavier particles. 

The Washington State Department of Transportation designed and constructed a scale 

model detention basin to investigate contaminant-removal capabilities. Using typical 

contaminants and concentrations, a simulated highway stormwater runoff was 

formulated and applied to the model detention basin over a range of flow rates. Four 

coagulants were evaluated for their ability to enhance removal of sediment and metals. 

Coagulant addition resulted in significant increases in metals removal over the range of 

stormwater flow rates studied. The greatest improvement was observed at the higher 

flow rates. Further improvements in contaminant removal were observed following the 

addition of an influent baffle. This baffle increased the hydraulic detention time by 

reducing short-circuiting with an associated improvement in contaminant removal (Price 

and Yonge, 1995). 

A restoration project began in 1987 in an effort to improve water quality in Lake Ella, 

Tallahassee, Florida. This effort constituted the first use of alum for treatment of 

stormwater inputs into a receiving water body. Lake Ella is a shallow, 5.3 ha (13 acre) 

lake that received stormwater runoff through 18 separate storm sewers from 64.8 ha 

(160 acre) of mostly impervious urban watershed. With a volume of 113,550 m3

 (30 

MG), Lake Ella receives 518,550 m3 (137 MG) of runoff each year. The lake was highly 

eutrophic and fish kills were common before installation of the alum stormwater 

treatment system. Initially, common stormwater treatment technologies such as 

retention basins and exfiltration trenches were considered. Alum treatment was 

considered when it was determined that there was no available land surrounding the 

lake that could be used for retention and the cost of purchasing homes to acquire land 

15
 
for construction of retention basins was cost prohibitive. The stormwater treatment 

system was implemented using flow meters and variable speed injection pumps to 

automatically inject liquid alum into the storm sewers. Mixing of the alum and 

stormwater occurs as a result of turbulence in the storm sewer line, and the floc 

produced settles on the lake bottom providing an added benefit of nutrient inactivation in 

the sediments. In addition to the non-structural alum retrofit, bar type trash traps were 

installed in underground structures upstream of the lake; the primary outfall lines were 

extended into the lake about 30.5 m (100 ft). The total cost of the alum injection 

system, excluding the extension of the storm sewer lines into the lake, was about 

$200,400. The capital cost of this system was $3,140/ha ($1,270/acre) of watershed 

area treated. The capital costs involved in constructing an alum treatment system are 

relatively independent of watershed size and dependent primarily upon the number of 

outfall locations treated (Harper and Herr, 1992). 

Polymer injection has also been used for the control of sewer overflows in the city of 

Dallas, Texas. An EPA-sponsored research program studied a non-structural retrofit 

system wherein the capacity of a sewer was increased by the injection of certain watersoluble chemicals to reduce turbulent friction. A polymer injection rate of 0.54 kg/min 

(1.2 lbs/min) into a 38 cm (15 in.), surcharged pipe conveying 18.9 m3

/min (5,000 

gallons per minute [gpm]) reduced head from 2.5 m (8 ft) above the pipe to 1.8 m (6 ft) 

above the pipe. The capacity of a sewage lift station was also increased by injection of 

a slurry of friction-reducing polymer into the pump intake. Polymer doses of 120 and 

350 ppm were sufficient at increasing flow in a 15.2 cm (6 in.) force main by 35% and 

64%, respectively (Chandler and Lewis, 1977). 

Real-Time Control 

Real-time control systems are gaining a renewed interest for the control of wet-weather 

overflows. In 1967, the U.S. federal government funded demonstration projects to 

investigate available technologies for control of CSOs. Demonstration grants were 

awarded to sewer agencies in Cleveland, Ohio; Seattle, Washington; Minneapolis, 

Minnesota; and Detroit, Michigan (Chantrill, 1990). These projects used computers for 

the coordination of storage at different sites in a sewer collection system to optimize use 

of storage on a system wide basis. All of the projects experienced problems getting the 

central computers operational largely due to the limitation of computer technology at the 

time. Today, municipalities have many more options for implementing computer-based 

real-time control of wet-weather overflows. 

Delattre (1990) outlined real-time control of combined sewer systems from the user’s 

perspective. Delattre’s discussion included design objectives and operating functions, 

implementation of supervision and control schemes, and evaluation of operation 

performance thorough case studies. Chantrill (1990) discussed the history of real-time 

control of combined sewers in the United States. Chantrill’s discussion included the 

control software developed in Seattle, Washington and Lima, Ohio, and the 

development efforts by Cello Vitasovic and Wolfgang Schilling. 

16
 
Treatment Facility Retrofits 

Increasing Capacity/Efficiency of Existing WWTF 

Retrofitting dry-weather treatment plants to handle increased flows from wet-weather 

events has proven to have merit in some communities. Typically it involves increasing 

the capacity of the existing treatment plant by installing newer, more efficient 

equipment. 

The town of Dexter, New York, solved an infiltration problem by upgrading a treatment 

plant with new aeration units rather than rehabilitating the collection system. Peak flows 

of more than 2,271 m3/d (0.6 MGD) from wet-weather events and groundwater 

infiltration were washing bacteria out of the 379 m3

/d (0.1 MGD) activated sludge 

WWTF. The town of Dexter decided to treat the excess infiltration, but did not have 

enough space at the plant to build a larger mixed-liquor system to handle excess flows. 

Installing prefabricated fixed-film treatment units that offered improved treatment 

efficiency and quick bacterial growth (Myers and Wickham, 1994) solved this problem. 

The city of Clatskanie, Oregon, conducted an infiltration/inflow (I/I) study and found that 

the collection system was subjected to severe I/I. Additionally, deficiencies existed in 

the trickling filter plant. A cost-effectiveness analysis indicated that the development of 

a joint dry/wet-weather facility without major collection system improvements was the 

best solution. The principal features of this alternative included a single primary clarifier 

followed by an activated sludge secondary treatment process with a DWF capacity of 

1,893 m3

/d (0.5 MGD). The secondary process can be operated in a contactstabilization mode, and the primary clarifier can be operated in a dissolved air flotation 

(DAF) mode. With these process modifications, flows up to 4,731 m3/d (1.25 MGD) 

could be effectively treated. SS and BOD removals during DWF were both 94%. WWF 

removal efficiencies for flows ranging from 1,893—8,706 m3/d (0.5—2.3 MGD) were 

71% and 73% for SS and BOD respectively. The capital cost of the DAF-contact 

stabilization capability was estimated to be 14% more than the cost of a standard dryweather plant (Whitney-Jacobsen and Associates, 1981). 

Modifications to the clarification system at the main WWTF in New Orleans, Louisiana 

have resolved a recurring hydraulic overloading problem caused by inflow and 

infiltration. The modifications also have increased average waste activated sludge 

concentrations by 50%. Sluice gates were installed to direct flows equally to the 

clarifiers during normal flows. These gates are adjusted during high flows to allow for 

optimal flows through the clarifiers, thereby eliminating sludge-blanket scouring that 

occurred before the gates were installed. A polymer injection system was also added to 

the thickener-clarifier to further enhance settling (Berry, 1994). 

The Acheres WWTF in Paris, France was designed to treat the carbonaceous portion of 

dry-weather sewage. The city is currently undergoing a major rehabilitation and 

retrofitting plan to upgrade the plant to include nutrient removal and wet-weather 

treatment. By the year 2001, the plant will be retrofitted with tertiary treatment for 

17
 
nutrient removal and physical-chemical treatment for wet-weather treatment. The wetweather treatment facility was designed to treat 0.76 m3

/min (200 gpm), almost twice 

the 0.42 m3/min (110 gpm) DWF (Gousailles, 1995). 

Physical-Chemical Treatment 

The physical-chemical treatment system at the Acheres wastewater plant will use the 

ACTIFLO process. This process incorporates lamellar settling and weighted 

flocculation by microsand. Guibelin, Delsalle and Binot (1994) presented operational 

and performance results from the ACTIFLO process. They established that the process 

allows an upflow velocity of 131 m/h (0.12 fps) over the lamella. Under this flow regime, 

80% SS and 60% BOD removal is achieved using 60g/m3 of ferric chloride and 0.8g/m3 

of polyelectrolyte. The process is also noted for its quick start up time, an important 

criteria for treating WWF. 

Another physical-chemical treatment system being studied in France is the Densadeg 

unit (Le Poder and Binot, circa 1995; Rovel and Mitchell circa 1995). This unit 

combines three principles: lamellar settling, coagulation-flocculation, and built-in 

thickening. The Densadeg pilot unit has been studied at WWTF and at representative 

sites along a combined sewer system. The researchers documented SS removals of 

80% at a velocity of 50 m/h (0.5 fps) and noted that coagulant dose (FeCl3) affected 

treatment results more than velocity. Densadeg units are to be installed at the new 

Colombes sewage treatment plant (240,000 m3/d [63.4 MGD]) in Paris, France. A key 

feature of the Colombes is how it will be operated during wet-weather events. Rather 

than sequential operation of the biological stage as during dry-weather, the stages will 

switch to operate in parallel during periods of high flow (Hayward, 1996). 

Nenov (1995) compared the effectiveness and costs of physical-chemical treatment 

versus the activated sludge process. Laboratory experiments confirmed the high 

efficiency of physical-chemical treatment for SS and BOD/COD removal. The data 

obtained show that chemical treatment over a large range of surface flow rate provides 

a reduction of SS and BOD/COD greater than 60% and 50%, respectively. Based on 

capital and O&M costs from the Ravad WWTF it was shown that physical-chemical 

treatment retrofitting could be done at a much lower cost than biological treatment 

expansion. 

Chemical Enhanced Treatment 

As an alternative to physical-chemical treatment, other municipalities have 

experimented with chemically enhanced treatment to increase plant capacity during 

wet-weather events. 

Sarnia, Ontario, studied the effectiveness of chemically enhanced primary treatment to 

increase and sustain removal efficiencies during high plant flows for more than ten 

years at an existing primary plant. Very little decrease in treatment efficiency was 

observed at overflow rates as high as 114.1 m3/d/m2

 (2,800 gpd/ft2). The study 

18
 
concluded that, with the addition of 14 ppm of ferric chloride and 0.3 ppm polymer, the 

SS and BOD removals averaged 84 and 60%, respectively (Heinke, Tay, and Qazi, 

1980). 

In March 1990, a 22 day test of chemically enhanced treatment was conducted at the 

South Essex Sewerage District (SESD) primary treatment plant in Salem, 

Massachusetts. The testing conditions were unique in that the flow was split and the 

plant was operated as two distinct and parallel treatment lines: half of the plant received 

chemical additives while the other half operated as a conventional primary plant. BOD 

removals on the conventional primary side averaged 17% but were 51% on the 

chemically treated side (a 200% increase). SS, fats, oils and grease removals 

increased by 60%. The test also demonstrated chemical addition can enable settling 

tanks to handle maximum stormwater conditions without deterioration in removal 

efficiency. Ferric chloride and polymer concentrations totaled approximately 30 ppm. 

Average overflow rates of 32.6 m3

/d/m2 (800 gpd/ft2) reached a peak of 77.4 m3/d/m2 

(1,900 gpd/ft2) during wet-weather events. The SESD plant's average DWF of 94,625 

m3

/d (25 MGD) increased to over 340,650 m3/d (90 MGD) during a storm on the ninth 

day of the test. SS and BOD removal rates did not decrease during or after this event 

(Harleman, Morrissey, Murcott, 1991). 

Converting Dry-Weather WWTF to a Wet-Weather Facility 

Conversion of existing or abandoned dry-weather treatment plants into wet-weather 

holding or treatment facilities is an option when communities build new dry-weather 

treatment plants. 

The Municipality of Toronto Metropolitan (Metro) has recently completed the Main 

Treatment Plant (MTP) Environmental Assessment (EA). One of the goals of the EA 

was to establish a preferred alternative for meeting CSO treatment needs. Metro 

placed a high priority on source control as a means of reducing CSO’s and associated 

pollutant loadings. Sewer separation was not considered as the only solution to the 

CSO problem because of concerns relating to increased stormwater discharges, the 

legality of separation on private properties, and high costs. The preferred strategy 

included, among other solutions, conversion of a WWTF to a CSO treatment plant 

(CG&S, 1997). 

4.2 Use of SWMM 

SWMM is a large, complex model capable of simulating the movement of precipitation 

and pollutants from the ground surface through pipe and channel networks, 

storage/treatment networks, and finally to receiving waters. The model is capable of 

simulating a single event or on a continuous basis for extended periods of time. 

SWMM has been released under several different official versions (Metcalf and Eddy, 

Inc., et al. 1971; Huber et al. 1975, 1984; Roesner et al. 1984; Huber and Dickinson 

19
 
1988; Roesner et al. 1988). The official versions were primarily designed for mainframe 

computer use; however, later versions have been modified for use on personal 

computers. The current “official” version, version 4, is available on disks or through the 

Internet from the EPA at the following addressees: 

Center for Exposure Assessment Modeling (CEAM)
 

U.S. Environmental Protection Agency
 

960 College Station Road
 

Athens, Georgia 30613
 

Phone: 706/355-8400
 

E-mail: ceam@epamail.epa.gov
 

WWW: ftp://ftp.epa.gov/epa_ceam/wwwhtml/ceamhome.htm
 

The complete SWMM model documentation (Huber and Dickinson 1988; Roesner et al. 

1988) can be obtained from the National Technical Information Services (NTIS) or 

Oregon State University: 

National Technical Information Service
 

5285 Port Royal Road
 

Springfield, Virginia 22161
 

Phone: 703/487-4650
 

NTIS no. PB88 236 641 – Huber and Dickinson (1988)
 

NTIS no. PB88 236 658 – Roesner et al. (1988)
 

Dr. Wayne C. Huber
 

Department of Civil Engineering
 

Oregon State University
 

Corvallis, Oregon 97331
 

Phone: 503/737-4934
 

E-mail: huberw@ccmail.orst.edu
 

The SWMM consists of five basic blocks (or components) that can be used together or 

separately. The blocks are RUNOFF, TRANSPORT, EXTENDED TRANSPORT 

(EXTRAN), STORAGE TREATMENT (S/T) and RECEIVING WATER (RECEIVE). The 

RUNOFF Block simulates rainfall, the resulting surface runoff quantities, and 

hydrographs for each drainage basin of the combined sewer collection system. The 

EXTRAN Block routes the surface runoff hydrographs developed in the RUNOFF Block 

through the collection system to the point of interception and overflow. 

SWMM was used to perform simplified simulations using historical rainfall data for a 

period of one year or the average of multiple years of hourly rainfall data with hourly 

time steps or an appropriate equivalent. This was done using the RUNOFF and 

TRANSPORT Blocks of SWMM. The RUNOFF Block generates surface runoff 

hydrographs from a drainage area in response to precipitation based on watershed 

characteristics including size, shape, imperviousness, and slope. The RUNOFF block 

20
 
uses the continuity equation that tracks the volume and depth of water on the ground 

surface and Manning’s equation that governs the rate of surface runoff. 

These hydrographs were then used as input into the TRANSPORT Block that routes 

flow through a sewer system that is depicted as a series of manholes, conduits, pump 

stations, storage tanks, and overflow structures. TRANSPORT uses a simplified 

version of the momentum equation where flow is a function of depth to route flow 

through the collection system. 

Unless otherwise noted, simplified simulations conducted for these analyses were done 

using existing calibrated and validated RUNOFF and TRANSPORT models, modified to 

estimate the pre- and post-retrofit conditions to assess the treatment/capture 

effectiveness of the retrofit. 

4.3 Case Studies and Hypothetical Retrofit Examples 

The desktop analyses provided in this document are a mixture of case studies of actual 

retrofit examples and hypothetical retrofit examples where actual retrofit examples were 

not readily available. The case studies include site specific details of each retrofit 

including description of facilities, operational parameters, flow rates, and costs. The 

hypothetical examples are used to provide a conceptual overview of retrofitting 

opportunities and allow comparison to other retrofitting examples. In some instances, 

actual case studies were labeled as “hypothetical” to allow the use of information 

without revealing the source to protect confidential information. In these instances, 

actual information was used to the greatest extent possible to increase the accuracy of 

the desktop analysis. 

21
 
Section 5
 

Desktop Evaluation
 

5.1 Conversion of Primary Treatment Tanks 

Conversion or retrofitting primary settling tanks at POTWs to provide equivalent or higher 

levels of treatment may be a technically feasible and economically attractive alternative to 

the construction of new conventional primary settling tanks. The principal downside of 

conventional primary treatment facilities is the large footprint required due to the low 

surface overflow rate (SOR). Alternative sedimentation methods including micro carrier 

enhancement, chemical precipitation, dissolved air floatation, and plate/tube settling may 

provide an alternative to conventional primary treatment. 

Two desktop analyses are presented to demonstrate the retrofit of primary treatment tanks 

to provide equivalent or higher levels of treatment: 

1. Retrofit of the Augusta Wastewater Treatment Facility, Augusta, ME 

2. Hypothetical retrofit of the POTW using the ACTIFLO process 

5.1.a. Augusta Wastewater Treatment Facility, Augusta, Maine 

The Augusta, Maine Sanitary District (ASD) owns and operates a secondary wastewater 

treatment facility (WWTF) that discharges to the Kennebec River, which is a Class C 

waterbody for the state of Maine. The WWTF was originally constructed in 1962 with 

upgrades in 1966, 1982 and 1997. The ASD WWTF provides secondary treatment using 

pure-oxygen activated sludge. The WWTF was designed to handle a monthly-average flow 

of 30,280 m3

/d (8 MGD) and an instantaneous-peak flow of 60,560 m3/d (16 MGD). The 

WWTF serves a partially combined collection system (CCS). ASD’s existing collection 

system services approximately 90% of the residential population, nearly all of the 

commercial properties, and a portion of the industrial properties within the limits of the city 

of Augusta. The system also serves four neighboring communities, the city of Hallowell and 

the towns of Manchester, Winthrop and Monmouth. The system contains three types of 

sewers: 

• combined sanitary and storms 

• separate sanitary 

• storm 

The majority of the collection system within the limits of Augusta flows by gravity to the 

WWTF. Ten pumping stations exist serving the lowlands and outlying areas. Currently, 

there are over 167 km (104 miles) of gravity sewer ranging from 15—107 cm (6—42 in.) in 

diameter. Pipe materials include stone and brick (dating back to the Civil War era) as well 

as more recent installations of vitrified clay, asbestos-cement, concrete, cast iron, and 

PVC pipe. According to the 1993 CSO Facilities Plan there are presently 38 diversion 

22
 
3 

manholes and 29 permitted CSO located within the system. In an average year 219,530 

m (58 MG) are discharged. 

A majority of the wastewater currently entering the ASD sewer system is domestic sewage 

from residential and commercial sources. The original WWTF design included an 

allowance for a significant organic loading from local industry; however, many of the local 

industries closed or relocated soon after Secondary Treatment System upgrade became 

operational in the early 1980’s. Consequently, the WWTF is operating with less than onehalf of its original design flow and organic loading. Currently, annual average wastewater 

flows are 15,140 m3

/d (4 MGD). 

Prior to implementation of the CSO Abatement Program, the WWTF used the following unit 

processes to provide preliminary, primary and secondary treatment: 

•	 One mechanically-cleaned bar screen with 3.8 cm (11⁄2 in.) clear spacing located in a 

1.5 m (57 in.) wide channel 

• One detritor-type grit chamber 4.3 m (14 ft) in diameter 

• Two 91.5 cm (36 in.) comminutors with a manual bypass screen 

•	 Three primary clarifiers, of which two are 16.8 m (55 ft) in diameter and one is 24.4 

m (80 ft) in diameter, for a combined surface area of 910 m2

 (9,800 ft2) 

• Four main primary effluent lift pumps 

•	 Two pure-oxygen activated sludge aeration tanks with a total volume of 2,725 m3 

(0.72 MG) 

•	 Three secondary clarifiers, each at 24.4 m (80 ft) diameter with a total surface area 

of approximately 1400 m2

 (15,100 ft2.) 

• One chlorine contact tank with a volume of 613 m3

 (162,000 gal) 

Figure 5.1.a-1 shows the existing site layout prior to implementing the CSO Abatement 

Program. Figure 5.1.a-2 shows a schematic of the existing WWTF prior to retrofitting. 

When influent peak flow approaches 34,065 m3

/d (9 MGD), excess flow is bypassed to the 

chlorine contact tank following degritting in the detritor. The bypass was automatically 

activated by a side overflow weir located along the grit chamber outlet channel. The water 

level was regulated by a 30.5 cm (12 in.) Parshall Flume located downstream of the weir. 

The CSO Abatement Program approved by the Maine Department of Environmental 

Protection recommended a four-phase plan to address CSO discharges. Phase I of the 

CSO Abatement Program included improvements to provide treatment for WWTF bypass 

that is the third largest CSO in the CCS based on projected peak flows. The following is a 

summary of Phase I Improvements, which is a retrofit to the existing facilities: 

23
 
Figure 5.1.a-1. Existing Site Layout Prior to Retrofit 

24
 
Figure 5.1.a-2. Process Flow Schematic Prior to Retrofit 

•
 

25
 
3

•	 A new headworks containing two mechanical screens with 2.5 cm (1 in.) clear 

spacing located in a 1.5 m (60 in.) wide channels, two 167 m3

 (44,000 gal) aerated 

grit chambers, and one 0.9 m (36 in.) parshall flume. The mechanically cleaned bar 

screens have screenings compactors to remove excess water. The grit is 

removed from the aerated grit chambers using recessed impeller type pumps 

which discharge to cyclone classifiers for washing and dewatering. 

•	 A new flow distribution structure (FDS No. 1) for the primary clarifier influent. The 

distribution structure utilizes motorized weir gates that will allow off-line tanks to be 

activated during a wet weather event. 

•	 A new primary effluent distribution structure (FDS No. 2) with motorized weir gate 

to regulate flow receiving secondary treatment and flow to the CSO bypass. 

•	 A new 250 m3

 (66,000 gal) chlorine contact tank with mechanical mixers using 

sodium hypochlorite for high-rate disinfection. A 24 m3

 (6,400 gal) bulk storage 

tank is provided for sodium hypochlorite storage. An 11 m3

 (3,000 gal) 

dechlorination contact chamber with mechanical flash mixer is provided. A 16 m3 

(4,150 gal) bulk storage tank is provided for sodium bisulfite. Chemical metering 

pumps are provided with flow pacing from an ultrasonic flow meter. 

These improvements will allow the full 109,800 m3

/d (29 MGD) of connected interceptor 

capacity to receive preliminary and primary treatment. Daily-maximum flows of 45,400 

m /d (12 MGD) would receive secondary treatment while 64,300 m3

/d (17 MGD) would be 

diverted to the high-rate disinfection system after receiving primary treatment. 

Later phases of the CSO abatement program (Phases II, III and IV) will allow additional 

WWF from the Augusta CCS to be treated at the WWTF. This will be accomplished 

through the construction of: 

•	 Consolidation conduits to intercept CSO discharges throughout the system and 

convey the flow to the WWTF; 

•	 Three new vortex separators, each 10.7 m (35 ft) diameter, and 2.4 m (8 ft) side 

water depth to provide the equivalent of primary treatment; 

•	 Additional high-rate disinfection and dechlorination tank capacity to increase the 

tank volume to 1,250 m3

 (330,000 gal) for disinfection, and 62.5 m3 (16,500 gal) for 

dechlorination; 

• A new wet-weather outfall from the WWTF to the Kennebec River. 

After all phases are completed, the total influent peak flow capacity to the WWTF would be 

increased from 109,800 m3

/d (29 MGD) to 405,000 m3/d (107 MGD). Up to 295,200 m3/d 

(78 MGD) will be treated in the vortex separators. The remaining 109,800 m3

/d (29 MGD) 

would be conveyed to the upgraded headworks. Approximately 268,700 m3

/d (71 MGD) of 

vortex separator effluent will be disinfected and dechlorinated in the expanded high-rate 

26
 
disinfection tank. Approximately 26,495 m3

/d (7 MGD) of vortex separator underflow would 

be conveyed to the upgraded headworks. A total flow of 136,260 m3

/d (36 MGD) would be 

treated within the headworks and primary clarifiers (109,765 m3

/d or 29 MGD influent plus 

26,495 m3

/d or 7 MGD vortex underflow). Primary clarifier effluent flows of approximately 

45,420 m3

/d (12 MGD) would receive secondary treatment. The remaining 90,840 m3/d 

(24 MGD) of primary clarifier effluent will receive high-rate disinfection and dechlorination. 

The basis of design of these facilities is based upon SWMM results using a one-year, twohour storm event. 

Figure 5.1.a-3 shows a site layout following completion of the CSO Abatement Program. 

Figure 5.1.a-4 shows a schematic of the ASD WWTF following completion of the CSO 

Abatement Program. 

It should be noted that the 1993 CSO Facilities Plan is currently being updated. The 

updated plan will re-assess the earlier recommendations for the later phases of the 

program. 

Stormwater Management Model Results 

The EPA SWMM Version 4 was used to assess the net benefit of the retrofit by comparing 

annual WWF volume treated at the WWTF under pre- and post-retrofit conditions. A 

simplified model was developed to simulate hourly rainfall, based upon the previously 

calibrated model developed for individual rainfall events. 

The data used for these projections consists of 21 years of hourly rainfall data from the 

National Weather Service (NWS) station in Portland, Maine for the years from 1971— 

1991. These data were used as input to the simplified RUNOFF model that projected 

flows that were used to simulate long-term overflow using the TRANSPORT block of 

SWMM. The TRANSPORT model was constructed as a simplified network consisting of 

overflows and regulator pipes. The simplified model was used to project the annual volume 

of treated WWF at the WWTF. 

Based on this analysis, the average annual WWF conveyed to the treatment plant was 

3 found to be 582,890 m (154 MG). The layout of the pre-retrofit headworks allowed 

484,480 m3

 (128 MG) to receive treatment consisting of primary and secondary treatment 

followed by disinfection while 98,410 m3

 (26 MG) of WWF annually bypassed the treatment 

plant and only received disinfection prior to discharge. The retrofitted headworks allowed 

the entire 582,890 m3 (154 MG) to receive primary treatment and 484,480 m3

 (128 MG) 

received secondary treatment and disinfection. The remaining 98,410 m3

 (26 MG) that did 

not receive secondary treatment was bypassed to the high-rate disinfection system prior to 

discharge to the Kennebec River. 

27
 
Figure 5.1.a-3. Site Layout Following Retrofit 

28
 
Figure 5.1.a-4. Process Flow Schematic Following Retrofit 

29
 
--

--

--

--

--

Design Considerations 

The retrofitting of the ASD WWTF to handle increased WWF required consideration of 

several factors: 

• Pollutant and Hydraulic Loadings 

• Treatment Objectives 

• Operational Requirements 

• Chemical Storage and Feed Requirements 

• Sludge Handling Requirements 

• Operations and Maintenance Requirements 

• Construction Sequencing and Site Constraints 

Each of these considerations is described in this section. 

Pollutant And Hydraulic Loadings - Pollutant and hydraulic loadings were developed for 

the wet-weather treatment units at the ASD WWTF during the Facilities Planning stage. 

These values were refined and updated as the project progressed through the preliminary 

engineering and detailed design stages. The primary objective of the wet-weather 

treatment units is to remove floatable and settleable solids to promote effective 

disinfection. 

Table 5.1.a-1 shows the design criteria for the wet-weather treatment processes at the 

ASD WWTF. The primary focus of the Phase I improvements are to remove hydraulic 

restrictions at the WWTF preventing flows of 109,800 m3

/d (29 MGD) from reaching the 

existing primary clarifiers. As noted previously, the completion of Phases II, III and IV will 

provide equivalent primary treatment and disinfection for additional peak WWF of 

approximately 295,200 m3

/d (78 MGD). 

Table 5.1.a-1. Phase 1 CSO Abatement Program Wet-Weather Treatment Design 

Criteria 

Description and Criteria Phase I Phases II, III & IV 

Vortex Treatment 

Number of Units (each) 0 3 

Design Flow 0.30 m3

/d 78.0 MGD 

Diameter Each 10.67m 35.0 ft 

Loading Rate 1098 m 3/d /m2 27,000 gpd/ft2 

Overflow to 268,735 m 3/d 71.0 MGD 

High-Rate Disinfection 

Underflow to Headworks 26,495 m3/d 7.0 MGD 

Mechanical Screening 

No. of Bar Screens (ea.) 2 2 

Type Reciprocating Rake Reciprocating Rake 

Bar Size 0.95x5.1cm 3/8x2 in 0.95x5.1 cm 3/8x2 in 

30 
Bar Spacing 2.54 cm 1 in 2.54 cm 1 in 

Table 5.1.a-1. Continued 

Description and Criteria Phase I Phases II, III & IV 

Peak Influent Flow
 109,765 m3/d 29.0 MGD 109,765 m3/d 29.0 MGD 

Vortex Overflow
 0.0 26,495 m3/d 7.0 MGD 

Total Peak Flow
 109,765 m3/d 29.0 MGD 136,260 m3/d 36.0 MGD 

Approach Velocity
 0.76 m/s 2.5 fps 0.84 m/s 2.75 fps 

Half Clogged
 

Screen Velocity
 2.07 m/s 6.8 fps 2.29 m/s 7.5 fps 

Half Clogged
 

Differential Head
 27.43 cm 0.9 ft 33.22 cm 1.09 ft 

Estimated Wet Screening
 3.7x10-5 m 3/m3 (5 ft3 /MG) 3.7x10-5 m 3/m3 (5 ft3 /MG) 

Quantity
 

Average De-watered
 0.28 m3/d 0.37 cyd/d 0.57 m3/d 0.74 cyd/d 

Screenings Volume
 

Screenings Container Size
 .57 m3/d 0.75 cyd 0.57 m3/d 0.75 cyd 

Screening Storage
 2 1 

Time (days)
 

Grit Removal
 

No. of Tanks
 2
 2 

Type
 Aerated
 Aerated 

Average Flow
 15,140 m3

/d 4.0 MGD
 30,280 m3/d 8.0 MGD 

Peak Flow
 109,765 m3

/d 29.0 MGD
 136,260 m3/d 36.0 MGD 

Volume Each Tank
 167 m3 44,000 gal 167 m3 44,000 gal 

Detention Time -
 31.7 16.8 

Avg. Flow (min)
 

Detention Time -
 4.4 3.5 

Peak Flow (min)
 

Air Supply
 0.27 to 3 to 8 cfm/ft 0.27 to 3 to 8 cfm/ft 

0.72 m3

/min/m 0.72 m3/min/m 

No. of Air Blowers
 2+1 standby 2+1 standby 

Air Flow Each
 2.4 to 84 to 224 cfm 2.4 to 84 to 224 cfm 

6.3 m3

/min 6.3 m3/min 

Pressure
 3867 kg/ m2 5.5 psig 3867 kg/ m2 5.5 psig 

Blower Motor Hp
 5/10, two speed 5/10, two speed 

Grit De-watering & Storage
 

Type
 Cyclone Cyclone 

No. of Classifiers
 2 2 

Flow
 1 m3/min 260 gpm 1 m3/min 260 gpm 

No. of Grit Pumps
 2+1 standby 2+1 standby 

Pump Head
 8.5 m 28 ft 8.5 m 28 ft 

Pump Speed (rpm)
 900 900 

Pump power
 7.5 kW (10 HP) 7.5 kW (10 HP) 

Grit Washer Diameter
 30.5 cm 12 in 30.5 cm 12 in 

Grit Washer Speed (rpm)
 18 18 

Estimated Grit Quantity
 45 cm3/m3 6.0 ft3/MG 45 cm3/m3 6.0 ft3/MG 

Grit Container Size
 4.6 m3 6 cyd 4.6 m3 6 cyd 

Grit Storage Time (days)
 7 3 

Description and Criteria
 Phase I Phases II, III & IV 

Primary Clarifiers
 

Average Flow 15,140 m3/d 4.0 MGD 30,280 m3/d 8.0 MGD
 

31 
Peak Flow
 109,765 m3

/d 29.0 MGD 136,260 m3/d 36.0 MGD 

BOD Loading 1
 3220 kg/d 7,100 lbs/d 7258 kg/d 16,000 lbs/d 

SS Loading 1
 5443 kg/d 12,000 lbs/d 7348 kg/d 16,200 lbs/d 

No. Of Tanks (ea.)
 3 3 

Table 5.1.a-1 Continued 

Description and Criteria Phase I Phases II, III & IV 

Diameter Each (ft) 2 @ 16.7 m 2 @ 55 ft 2 @ 16.7 m 2 @ 55 ft 

1 @ 24.4 m 1 @ 80 ft 1 @ 24.4 m 1 @ 80 ft 

Sidewater Depth
 2.1 m 7.0 ft 2.1 m 7.0 ft 

Total Volume (MG)
 1938 m 3 0.512 1938 m 3 0.512 

Average Hydraulic Loading
 16.7 m3/d/ m2 409 gpd/ft2 33.3 m3/d/ m2 818 gpd/ft2 

Peak Hydraulic Loading
 121 m3/d/m2 2,966 gpd/ft2 150 m3/d/ m2 3,682 gpd/ft2 

Average Weir Loading
 83.2 m3/d/m 6,701 gpd/ft 166.5 m3/d/ m 13,403 gpd/ft 

Peak Weir Loading
 603.4 m3/d/m 48,584 gpd/ft 749.0 m3/d/ m 60,311 gpd/ft 

Detention Time –
 3.1 1.5 

Average Flow (h)
 

Detention Time –
 0.4 0.3 

Peak Flow (h)
 

Estimated BOD5 Removal1
 1352 kg/d 2,980 lbs/d 2205 kg/d 4,860 lbs/d 

Estimated SS Removal1
 2994 kg/d 6,600 lbs/d 3674 kg/d 8,100 lbs/d 

Estimated Sludge
 40,000 40,000 

Concentration (mg/L)
 

Estimated Sludge
 75 m3/d 0.0198 MGD 92 m3/d 0.0243 MGD 

Quantity @4.0%1
 

Primary Sludge/Scum
 5 5 

Pumps
 

Pump Type
 Disc Disc 

Pump Flow Each
 0.3 m3/min 85 gpm
 0.3 m3/min 85 gpm 

Total Head
 4.6 m 15 ft
 4.6 m 15 ft 

Pump Speed (RPM)
 300
 300 

Pump power
 3.7 kW (5 HP)
 3.7 kW (5 HP) 

No. of Grinders
 5
 5 

CSO Disinfection 

Chemical Type
 Sodium Hypochlorite
 Sodium Hypochlorite 

Peak Flow
 64,345 m3/d 17.0 MGD
 359,575 m3/d 95.0 MGD 

Dose Rate at Peak Flow
 15.0
 15.0 

(mg/L)
 

Chemical Concentration
 15.0 15.0 

(%)
 

Cl2 Feed Rate
 149.8 kg/m3 1.25 lbs/gal 149.8 kg/m3 1.25 lbs/gal 

No. of Pumps
 1 + 1 standby 2 + 1 standby 

Pump Capacity each
 0.27 m3/h 70.0 gph 0.76 m3/h 200.0 gph 

Contact Time (min)
 5.5 5.0
 

Contact Chamber Volume
 250 m3 66,000 gal 1250 m 3 330,000 gal
 

Length to Width Ratio
 7:1 40:1
 

Description and Criteria
 Phase I Phases II, III & IV
 

CSO Dechlorination 

Chemical Type Sodium Bisulfite Sodium Bisulfite 

Peak Flow 64,345 m3

/d 17.0 MGD 359,575 m3/d 95.0 MGD 

Chlorine Residual (mg/L) 5.0 5.0 

NaHSO3 Dose (mg/L) 8.06 8.0 

32
 
NaHSO3, Concentration (%) 38.0 38.0
 

Feed Rate NaHSO3 504.4 kg/m3 4.21 lbs/gal 504.4 kg/m3 4.21 lbs/gal
 

No. of Pumps 1 + 1 standby 2+1 standby
 

Pump Capacity each 0.05 m3/h 12.0 gph 0.12 m3/h 32.0 gph
 

Contact Time (s.) 15.0 15.0
 

Contact Chamber Volume 11.4 m3 3,000 gal 62.5 m3 16,500 gal
 

Treatment Objectives. The state of Maine requires an Escherichia coliform (E-coli) 

concentration of 949 colonies per 100 ml to comply with the maximum daily E-coli limit in a 

Class C river. Sampling and monitoring performed during the development of the CSO 

Abatement Program indicated that the Kennebec River does not currently meet water 

quality criteria for bacteria during wet weather. The CSO Abatement Program 

recommended a plan to treat CSO discharges to meet the current maximum-daily-waterquality limit. Primary treatment and high-rate disinfection were selected as the most 

economical methods to meet the treatment objectives for this project. 

Operational Requirements. The Phase I Improvements to the Augusta WWTF designed 

to operate on a continuous basis include: 

• Mechanical Screening 

• Grit Removal 

• Primary Treatment 

The high-rate disinfection and dechlorination systems are intended to operate in an eventbased mode during periods of excessive WWF. Under current DWF conditions, the 

WWTF operates with only one primary clarifier in service. The other two primary clarifiers 

can be kept out of service and used as off-line storage during wet-weather events. If one 

24.4 m diameter (80 ft) primary clarifier is in service, the remaining two 16.8 m (55 ft) 

diameter primary clarifiers provide approximately 943 m3

 (249,000 gal) of off-line storage. 

The WWTF staff can activate the off-line tanks using the new Supervisory Control and Data 

Acquisition (SCADA) System Controls. Motor-operated weir gates regulate the flow to the 

primary clarifiers. 

After the WWF subsides the primary clarifiers can be drained using a new pump with 

variable speed controls. The primary clarifier tank drain pump discharge is returned to the 

new Primary Clarifier Flow Distribution Structure (FDS No. 1). 

The future improvements of Phases II, III and IV will use an event-based mode of operation 

during wet weather periods. The existing overflow pipes located within the CSS will be 

intercepted by new consolidation conduits that will convey the combined sewerage to the 

WWTF. A series of vortex separators will be constructed to provide equivalent primary 

treatment (removing floatable and settleable solids) to facilitate effective disinfection. The 

high-rate disinfection and dechlorination tank will be expanded with a new parallel outfall 

pipe provided to the Kennebec River from the WWTF. At the end of the wet-weather event 

the vortex separators and high-rate disinfection tank will be drained using a pump with 

variable speed control to the WWTF headworks. 

33
 
Chemical Storage and Feed Requirements. The CSO Abatement facilities will utilize two 

chemicals: 

• Liquid Sodium Hypochlorite for disinfection 

• Liquid Sodium Bisulfite for dechlorination 

The liquid sodium hypochlorite system design will use on commercial-grade liquid sodium 

hypochlorite (15% solution) which contains the equivalent of 150 kg of chlorine/m3

 (1.25 lbs 

of chlorine/gal of solution). The metering pumps are designed to deliver a dosage of 15 

milligrams per liter (mg/L) at the peak design flow with one unit out of service. A bulk 

storage tank with an effective working volume of 240.2 m3

 (6,400 gal) is provided. The 

tank is used for both dry-weather, secondary effluent disinfection, and high-rate wetweather disinfection to minimize problems associated with product shelf life. The available 

chlorine in sodium hypochlorite solution declines with age. The anticipated sodium 

hypochlorite storage time will be 51 days initially, decreasing to 25 days at the future 

design flow. 

The storage system design is based on commercial grade liquid sodium bisulfite (38% 

solution) which contains approximately 500 kg of sodium bisulfite per cubic meter of 

solution (4.2 lbs of sodium bisulfite/gal of solution). The metering pumps are designed to 

deliver sodium bisulfite dosages necessary to dechlorinate effluent with a chlorine residual 

of 5.0 mg/L. Approximately 1.6 kg of sodium bisulfite will be used per kg of chlorine 

residual. A bulk storage tank with an effective working volume of 15.7 m3

 (4,150 gal) is 

provided. 

The sodium hypochlorite and sodium bisulfite metering pumps are flow paced. An 

ultrasonic level sensor is used for effluent flow measurement. A broad-crested rectangular 

weir 3.4 m (11 ft) in length is the primary measuring device. 

Sludge Handling Requirements. Sludge handling requirements will increase during 

periods of WWF. The most pronounced increase is expected to occur during the initial 

period of WWF (first flush). The configuration of FDS No. 1 will capture a significant 

portion of the first flush solids in the off-line primary clarifiers. 

A solids balance was performed to estimate the quantity of sludge produced during a 

storm with a one-year recurrence interval. The daily sludge quantities increase as follows: 

• Current Estimated Sludge Quantity (kg/d) 3,290 (7,250 lbs/d) 

• One-Year Storm Estimated Sludge Quantity (kg/d) 7,550 (16,650 lbs/d) 

The sludge quantity is estimated to increase 230% during a one-year storm event. 

However, over the course of a week, the overall increase can be assimilated if sludge 

pumping and processing schedules are adjusted. The following is a summary of weekly 

sludge quantities during a period with a one-year storm: 

• Initial Estimated Sludge Quantity (kg/wk) 23,010 (50,730 lbs/wk) 

34
 
• One-Year Storm Projected Sludge Quantity (kg/wk) 27,270 (60,140 lbs/wk) 

The estimated sludge quantity would only increase by 19% during a week with a one-year 

storm event. 

Operation and Maintenance Requirements. The implementation of the CSO Abatement 

Program will result in increased O&M costs. The following list summarizes the areas of 

expected O&M cost increases at the WWTF: 

•	 Labor to perform sampling, laboratory analyses, reporting, wet-weather process 

adjustments, and clean-up following an event. 

• Power for in-plant pumping, sludge processing, and chemical feed systems. 

•	 Chemicals for disinfection, dechlorination, and polymer for additional sludge 

quantities. 

• Sludge disposal tipping and transport fees. 

Table 5.1.a-2 summarizes the expected increase in annual O&M costs at the WWTF 

resulting from the initial Phase I and future Phases II, III and IV of the CSO Abatement 

Program. 

Table 5.1.a-2. Estimated O&M Costs, CSO Abatement Program - All Phases 

Description Estimated Annual Cost 

Sampling
 $800 

Laboratory Analyses
 $3,200 

Report Preparation
 $1,500 

Process Control Adjustments
 $1,200 

Clean-Up of Wet-Weather Treatment Structures
 $6,000 

Power
 $1,000 

Sodium Hypochlorite
 $4,100 

Sodium Bisulfite
 $2,000 

Polymer for Sludge Dewatering
 $1,000 

Sludge Disposal Transport and Tipping Fees
 $11,500 

Total Estimated Cost $32,300 

Construction Sequencing and Site Constraints. The existing ASD WWTF site is 

crowded, with little room for expansion. Therefore, it was necessary to construct some of 

the CSO Abatement Facilities in locations previously used for other purposes. The 

following is a listing of special sequencing and site considerations related to the CSO 

Abatement Program at the WWTF: 

•	 Portions of the existing screen structure and detritor grit chamber were demolished 

to facilitate the construction of the new headworks. 

• A temporary junction structure and CSO bypass was needed. 

35
 
•	 The existing primary clarifiers were taken out of service in sequence, to facilitate 

new interconnecting piping and mechanical refurbishment. 

•	 The existing comminutor structure required temporary piping connections to keep it 

in operation during the new headworks construction. 

•	 After the new headworks are built and placed into service, the existing comminutor 

structure will require demolition and removal to facilitate installation of new 

underground piping. 

A completion time of 540 calendar days was specified for the Phase I CSO Abatement 

Program. 

Project Costs 

The bids for the Phase I CSO Abatement Program were opened on January 23, 1997. 

Costs for the CSO Abatement components are listed in Table 5.1.a-3. Costs included 

construction requirements and project services. Project services included design 

engineering, construction engineering, legal, and fiscal expenses but did not include 

contingencies, right-of-way expenses, or other miscellaneous items. 

Table 5.1.a-3. Phase 1 CSO Abatement Program Project Costs 

Description Cost1 

General Conditions
 

Sitework and Yard Piping
 

Headworks Building and Chemical Feed Systems
 

CSO Disinfection and Dechlorination Structure
 

Primary Pump Room Renovations
 

Primary Clarifier Renovations
 

Flow Distribution Structure No. 1
 

Flow Distribution Structure No. 2
 

Operations Building Pump Room Renovations
 

Yard Electrical Ductbanks and Lighting
 

Instrumentation and SCADA System
 

Total Construction Cost 

Project Services2 

$200,000 

$344,000 

$1,688,000 

$536,000 

$80,000 

$40,000 

$117,000 

$122,000 

$60,000 

$88,000 

$167,000 

$3,422,000 

$761,000 

Total Project Cost $4,183,000 1 Based on January 23, 1997 Bid Prices 

2 Includes Design Engineering, Construction Engineering, Legal, and Fiscal 

Expenses; does not include contingencies and right-of-way costs. 

Conclusions 

This retrofit will utilize the entire capacity of the existing primary clarifiers to provide primary 

treatment for the maximum rate of influent flow, thereby increasing the portion of the WWF 

that will receive primary treatment from 60,560 m3

/d (16 MGD) to 109,765 m3/d (29 MGD). 

This will result in net 20% increase: 484,480—582,890 m3

 (128—154 MG) in the annual 

36
 
3

WWF that receives primary treatment. Only a few examples of retrofitting WWTF have 

been published. Often these examples are strongly influenced by site-specific constraints 

and therefore do not lend themselves to comparison with other retrofit situations. The 

retrofit solution described here has been selected based on the cost effectiveness. For 

comparison purposes, the alternative treatment scheme of off-line storage at the WWTF for 

the 13 MGD of WWF was estimated to be $8,336,000 versus the $4,183,000 for the 

selected retrofit. 

5.1.b. Hypothetical Retrofit of Primary Treatment Facilities Using the ACTIFLO 

System 

ACTIFLO System 

This desktop analysis involves the retrofit of hypothetical primary treatment tanks for 

enhanced treatment of WWF in excess of the 908,400 m3

/d (240 MGD) peak wet-weather 

capacity of the existing WWTF. The treatment method selected for this retrofit scenario is 

the ACTIFLO process that combines microsand-enhanced flocculation with lamellar plate 

settling. The process has been shown to be effective in treating CSO/SSO and primary 

wastewater flows in installations in Paris, France (Le Poder and Binot 1994). Typical 

performance results are 85-95% reduction in SS, 50—80% reduction in Total BOD, 85— 

95% reduction of total phosphorus and 10—20% reduction in total kjeldahl nitrogen (TKN). 

The hypothetical WWTF has eight primary clarifiers that are 41.2 m (135 ft) in diameter 

with a 3 m (10 ft) side water depth. The surface-settling rate of the existing units at the 

3 3 design peak flow of 908,400 m/d (240 MGD) is 85.6 m/d/m2

 (2,100 gpd/ft2) with a 

hydraulic detention time of 48 min. Figure 5.1.b-1 shows the existing layout of the 

hypothetical WWTF prior to construction of the Phase I. 

Due to the limited side water depth, retrofit of the existing primary clarifiers is not practical. 

A more economical approach is the demolition of one of the existing clarifiers and 

replacing it with a 908,400 m3

/d (240 MGD) ACTIFLO system which has a depth of 6.7 m 

(22 ft). The seven remaining clarifiers would be retained for storage of WWF during storm 

events or for future plant modifications. 

The proposed 908,400 m3

/d (240 MGD) ACTIFLO system design consists of six 151,400 

m /d (40 MGD) process trains. Each of the process trains is designed to operate at a 

nominal overflow rate of 1.2 m3

/min/m2 (30 gpm/ft2) with a total hydraulic retention time of 

5.6 min. 

Table 5.1.b-1. ACTIFLO Design Summary 

Design Parameter Quantity 

Number of Process Trains 6 

Nominal Train Capacity	 151,400 m3

/d 40 MGD 

Total System Capacity 908,400 m3

/d 240 MGD 

37
 
Hydraulic Retention Time 5.6 min 

Figure 5.1.b-1. Layout of Hypothetical WWTF Prior to Construction of ACTIFLO 

System 

38
 
The high overflow rate and short retention time offered by the ACTIFLO process result in an 

extremely compact system. Overall, the proposed ACTILFO design will meet the treatment 

objectives of the project in approximately 5% of the space currently occupied by the 

existing primary clarifiers. A brief comparison of the existing facilities and the ACTIFLO 

design are shown in Table 5.1.b-2. 

Table 5.1.b-2. Comparison of ACTIFLO and Existing Primary Treatment Systems 

Net Savings with 

Parameter Existing System ACTIFLO ACTIFLO 

Total Capacity 908,400 m3

/d
 908,400 m3/d ­

Number of Units 8
 6 ­

Unit Capacity 113,550 m3

/d
 151,400 m3/d ­

Area Required 1,329.1 m2

 (per unit)
 87.8 m2 (per train) 526.7 94% per unit 

10,632.7 m2

 (total) m2 (total) 95% total 

Process Volume 3,987 m3

 (per unit) 3,533 m3 (per unit) 85% per unit 

31,898 m3

 (total) 3,588.3 m3 (total) 89% total 

Hydraulic Retention 51 min 5.6 min 89% 

Time 

A layout drawing of the hypothetical WWTF showing the location of the proposed ACTIFLO 

system is provided in Figure 5.1.b-2. 

ACTIFLO Process Theory 

Fundamentally, the ACTIFLO process for wastewater treatment is very similar to 

conventional water treatment technology of coagulation, flocculation and sedimentation. 

Although influent characteristics vary for different wastewater applications and influent 

streams, the removal of suspended materials is accomplished by the same mechanisms in 

each application. Therefore, for the purposes of this discussion, the term wastewater 

(WW) collectively refers to any domestic wastewater stream including primary sewage, 

CSO, SSO and/or inflow and infiltration (I/I). 

As in conventional water treatment technology, a coagulant is used for the destabilization of 

suspended materials entering the process and a flocculent aid polymer is added to 

aggregate the destabilized solids into larger masses. The resulting flocs are then 

subsequently removed by settling for disposal. The primary advance made in the 

ACTIFLO process is the addition of very fine sand as a “seed” for the development of highdensity floc. The resulting flocs, ballasted by the microsand, are more easily removed by 

settling. It is in this step that ACTIFLO differs significantly from conventional treatment 

processes. A brief overview of the physiochemical principals involved in conventional 

coagulation and flocculation is presented in "Section 5.2.b. Chemical Addition to AWPCP" 

to help understand the advantages of ACTIFLO over chemical coagulation alone. 

39
 
Figure 5.1.b-2. Layout of Proposed ACTIFLO System in Hypothetical Setting 

40
 
The primary advance made in the ACTIFLO process is the addition of a very fine (60—140 

�m mean diameter) clean silica sand or microsand in the flocculation process step. With 

its addition, the microsand serves several key roles in the ACTIFLO process: 

•	 The high specific surface area to volume ratio of the microsand particles serves as a 

seed for the development of floc. 

•	 Together, the microsand and polymer seed promote inter-particle contact and 

enmeshment of suspended materials resulting in development of large stable floc. 

•	 The high microsand concentration within the ACTIFLO process effectively dampens 

the effects of fluctuating influent quality and provides for stable operation and 

treatment throughout changes in the raw water quality. 

•	 The relatively high specific gravity of the microsand (~2.65) serves as a ballasting 

agent creating heavier floc that settle very quickly. 

•	 The chemically inert microsand does not react with the process chemistry and allows 

the microsand to be efficiently separated from the sludge via a hydrocyclone and 

reused in the process. 

The advantages of microsand ballasted flocculation allow ACTIFLO to offer surface loading 

or overflow rates many times higher than competing processes. Overflow rates as high as 

3.3 m3

/min/m2 (80 gpm/ft2) are claimed in typical wastewater applications of ACTIFLO. 

The high loading rates result in a process that is extremely compact compared to 

conventional or competing processes of similar capacity. The high overflow rates and size 

savings provided by ACTIFLO translate directly into significant savings in project cost. 

The use of microsand for ballasted flocculation provides ACTIFLO with several operational 

advantages in wastewater applications. The high solids content within the ACTIFLO 

process provides for effective treatment of influent with varying SS concentrations. In such 

applications, the high microsand concentration within the process allows for SS removal by 

particle enmeshment in the microsand-polymer complex. This performance is also 

extremely beneficial in chemical precipitation reactions such as tertiary phosphorus 

removal/effluent polishing or the treatment of dilute waste streams. 

Microsand ballasted flocculation is also effective in the treatment of extremely dirty 

wastewater (high turbidity and SS). Here, the high microsand concentration within the 

process allows it to provide efficient removal of high-suspended solids concentrations 

without difficulty. For similar reasons, the process is virtually unaffected by sudden 

fluctuations in influent quality or volume. Minor adjustments to chemical dosage are all that 

are needed to provide effective treatment under such conditions. 

Overall, the use of microsand results in the development of chemical floc that is significantly 

denser and more durable than floc from conventional clarification processes. The 

41
 
ACTIFLO flocs have considerably higher settling velocities than conventional floc allowing 

much higher clarifier loading rates. Ballasted flocculation also provides for a process that 

is rapidly started-up and optimized, as well as stable in treating with variations in raw water 

quality. These characteristics make the ACITLFO process ideal for treating the difficult 

conditions posed by wastewater/CSO treatment. 

ACTIFLO Process Description 

The ACTIFLO®

 process is a compact high performance water clarification system that 

combines the advantages of microsand enhanced flocculation with lamellar settling. 

Wastewater treatment is accomplished through a series of consecutive process steps that 

consist of influent screening, coagulant addition, microsand and polymer injection 

(injection), floc maturation (maturation), settling and sand recirculation. Each step in the 

process is discussed in greater detail in the paragraphs that follow. A flow 

diagram/process schematic of the ACTIFLO system is provided in Figure 5.1.b-3. 

Prior to entering the ACTIFLO process, raw wastewater is directed through fine screens to 

remove large debris and prevent equipment fouling in the process. Debris removed by the 

fine screening process is typically sent directly to a sanitary landfill for disposal. Chemical 

coagulant is then added to the screened raw water in the influent line via an in-line 

mechanical or jet mixing system to provide for thorough and instantaneous dispersion of 

the coagulant into the influent stream. The coagulant, typically ferric 

42
 
chloride (FeCl3) in wastewater applications, serves to destabilize suspended materials as 

they enter the process. 

The coagulated water enters the ACTIFLO system in the injection tank. Here, flocculent aid 

polymer (polymer) and microsand are added to initiate floc formation. Microsand and 

polymer are incorporated into the coagulated water via mixing over 1 min of retention time. 

Together, the microsand and polymer form ballasted pin-floc that is allowed to develop 

further in the next process step. 

Treatment continues as water passes through the underflow passage from the injection 

tank into the maturation tank. Although chemical floc formation actually begins with the 

addition of polymer and microsand in the injection tank, the majority of ballasted floc 

formation occurs during the maturation process step. Gentle mixing and increased 

hydraulic retention time of approximately 3 min provide ideal conditions for the formation of 

polymer bridges between the microsand and the destabilized SS. This process is further 

augmented by the large specific surface area of the microsand that provides enhanced 

opportunity for polymer bridging and enmeshment of suspended materials. 

Fully formed ballasted floc leave the maturation tank and enter the settling tank where they 

rapidly settle and are removed from the treated water via lamellar settling. Laminar up flow 

through the lamellar settling zone provides effective removal of smaller floc and suspended 

materials in the treated water. Clarified water exits the ACTIFLO system by a series of 

collection troughs for subsequent treatment. 

Sludge Handling. The ballasted floc sand-sludge mixture is collected at the bottom of the 

settling tank and withdrawn using a rubber-lined centrifugal slurry pump. The sand-sludge 

mixture is then pumped to the hydrocyclone for separation. Energy from pumping is 

effectively converted to centrifugal forces within the body of the hydrocyclone causing 

chemical sludge to be separated from the higher density microsand. Once separated, the 

microsand is concentrated and discharged from the bottom of the hydrocyclone and reinjected into the process for re-use. The lighter density sludge is discharged from the top 

of the hydrocyclone and sent for thickening and/or final disposal. 

Coagulant Process Data. As previously discussed, chemical coagulant is added to the 

raw water to destabilize suspended materials as they enter the ACTIFLO system. To 

provide effective treatment, coagulant should be injected in the influent line upstream of the 

system. In-line mechanical or jet injection mixing, depending on flow rate should be 

provided to ensure nearly instantaneous incorporation of the coagulant into the influent 

stream. 

Although any number of different coagulants could be used, FeCl3 is typically used in 

wastewater and CSO treatment applications since it is less expensive, and thus more 

economical in high capacity treatment applications than other available coagulants. 

FeCl3 also offers the advantages of being less sensitive to pH and more aggressive in the 

removal of organics. The only disadvantage to FeCl3 use is the need for corrosion 

43
 
resistant coagulant dosing and mixing equipment in the influent line. Once incorporated 

into the raw water, FeCl3 does not pose a serious corrosion threat to process equipment 

within the ACTIFLO system. 

Stormwater Management Model (SWMM) Results 

The EPA SWMM Version 4 was used to assess the benefit of the retrofit by comparing 

annual WWF volume treated at the WWTF under pre- and post-retrofit conditions for the 

hypothetical retrofit. A simplified model was developed to simulate hourly rainfall for the 

hypothetical collection system. 

The data used for these projections consists of 30 years of hourly rainfall data for the years 

from 1962—1991. These data were used as input to the simplified RUNOFF model that 

projected flows that were used to simulate long-term overflow using the TRANSPORT 

block of SWMM. The TRANSPORT model was constructed as a simplified network 

consisting of hypothetical overflows and regulator pipes. The simplified model was used to 

project the annual volume of treated WWF at the WWTF. 

Table 5.1.b-3 demonstrates that the hypothetical ACTIFLO system would provide 

considerable annual capture of WWF that is presently discharged without treatment. This 

is due to the use of the remaining seven primary treatment tanks as wet-weather storage 

devices with the ACTIFLO system. These tanks have a combined storage capacity of 

28,388 m3

 (7.5 MG). 

Table 5.1.b-3. Average Annual Overflow Volume from the Pre- and Post-Retrofit Facilities 

(1962 – 1991) 

Average Annual Overflow Volume 

Pre-Retrofit WWTF Overflow 439,060 m3

/d 116 MG 

Post-Retrofit 34,065 m3/d 9 MG 

Percent Capture 92% 

The proposed ACTIFLO system in combination with the remaining 7 primary clarifiers 

modified for storage of WWF would provide a system to treat 92% of the WWF that is 

presently discharged at the WWTF. 

Project Costs 

Operational Costs. The primary costs associated with the operation of the ACTIFLO 

system are electrical power and process chemicals. The major power components in the 

system are the mixers, scraper drive and microsand recirculation pumps. 

44
 
Additionally, a small amount of electrical power is required for the PLC control panels, 

polymer mix and feed system, and coagulant feed system. Wherever possible, energy 

efficient equipment is selected for use in the ACTIFLO system to help reduce overall 

operating costs. In wastewater/CSO treatment applications, the ACTIFLO system is 

typically operated at less than full capacity, further reducing electrical costs. 

Chemical costs are the other major expense associated with operation of the ACTIFLO 

process. The chemical costs consist of primarily of coagulant and flocculent aid polymer. 

Depending on the raw water characteristics, pH adjustment chemical costs might also be 

required in certain applications. As with any water treatment process, chemical dosages 

are difficult to predict without pilot or full-scale operating data from the site. Furthermore, 

chemical dosages are subject to vary with changes in the influent conditions. Therefore, a 

range of typical or anticipated chemical dosages is used to calculate operating costs. 

Although chemicals are a major component in the overall operating costs of the ACTIFLO 

process, the chemicals are used very efficiently due to the mixing provided within the 

system. Overall, ACTIFLO is typically capable of producing similar or better treatment 

results with 20—50% reductions in chemical usage as compared to conventional physicalchemical treatment systems of similar capacity. The estimated annual O&M costs are 

provided in Table 5.1.b-4. 

Capital Costs. Capital costs estimates for retrofitting the proposed ACTIFLO system into 

the existing WWTF that were provided by the ACTIFLO manufacturer (Kruger Inc.) are 

listed in Table 5.1.b-5. The costs include construction requirements and project services 

including design engineering, construction management, legal and fiscal expenses but do 

not include contingencies or other miscellaneous items. 

The conventional abatement alternative for the hypothetical retrofit would be to construct 

wet-weather storage facilities adjacent to the existing primary treatment facilities to 

equalize WWF for ultimate treatment at the WWTF. This would require the construction of 

a 28,388 m3

 (7.5 MG) overflow retention facility (ORF). The estimated cost to construct this 

facility based upon EPA cost guidelines for CSO control technologies (EPA 1992) is $18.4 

million derived from the following equation: 

Cost = 3.577 V0.812 (1)
 

where,
 

Cost = Average construction cost, millions of dollars
 

V = Storage volume, MG
 

45
 
Table 5. 1. b-4. Estimated O&M Costs 

DESCRIPTION Estimated Annual Cost 

SAND COMSUMPTION 

Est. Sand Loss (g/m 3 water 

produced) 

1 

Design Flow Rate (m 3/d) 908,400 (240 MGD) 

Sand Loss (kg/day) 793 (1,748 lbs/d) 

Sand Loss (mt/y) 263 (290 tons/y) 

Sand Cost ($/ton) $88 ($80/ton) 

Typical Operation Cost ($/y) (1) $9,253 

POLYMER COMSUMPTION 

Est. Polymer Dosage (ppm) 0.8 

Polymer Consumption (kg/d) 205 (452 lbs/d) 

Polymer Consumption (mt/y) 68 (75 tons/y) 

Polymer Cost ($/mt) $3,858 ($3,500/ton) 

Typical Operation Cost ($/y) $104,694 

COAGULANT COMSUMPTION 

Coagulant Type Ferric Chloride 

Est. Coagulant Dosage (ppm) 65 

Coagulant Consumption (kg/d) 26,008 (57,501 lbs/d) 

Coagulant Consumption (mt/y) 8,650 (8,540 tons/y) 

Coagulant Cost ($/mt) $220 ($200/ton) 

Typical Operation Cost ($/y) $760,978 

ENERGY COMSUMPTION 

Injection Mixer kW per Train 11 (15 HP) 

Maturation Mixer kW per Train 15 (20 HP) 

Scraper kW per Train 3.7 (5.0 HP) 

Sand Pump kW per Train 37 (50 HP) 

kW per Train 67 (90 HP) 

Total kW 403 (540 HP) 

kW/1,000 m3 0.44 (2.25 HP/MGD) 

KW/d 7,725 (9,688 HP/d) 

Typical Operation Cost ($/y) $107,090 

Total Typical Operation Cost 

($/y) 

$982,015 

(1) Typical annual O&M costs are based on operation at 33% capacity 302,800 m3/d (80 MGD) 

for 365 days per year with the provision for 75 wet-weather events lasting 12 h each requiring 

operation at 100% capacity. 75 wet-weather events per year are based upon long-term 

continuous simulation of 30 years of data through a hypothetical collection system using the 

TRANSPORT block of SWMM. 

46
 
3

Table 5.1.b-5. Estimated ACTIFLO Retrofit Capital Costs 

Description Estimated Cost(1) 

Equipment Cost 

Installation Cost 

Concrete Cost 

$12,000,000 

$960,000 

$700,000 

Total Project Cost $13,660,000 

(1) Includes Design Engineering, Construction Engineering, Legal, and 

Fiscal Expenses based on information provided by ACTIFLO 

manufacturer, Kruger, Inc. Does not include contingencies and right-ofway costs 

In comparison, the capital cost for the ACTIFLO process is less costly than the 

conventional storage facility, $14.5 million versus $18.4 million; however, the projected 

O&M cost of the ACTIFLO facility at almost $1 million per year make it a less cost-effective 

alternative to storage. This is a result of high chemical costs associated the ACTIFLO 

process. 

Typical ACTIFLO Performance in WW/CSO Treatment 

ACTIFLO Pilot Testing Results. A pilot plant with a nominal capacity of 150 m3

/h and a 

maximum capacity of 200 m3

/h was used by Kruger to conduct the pilot testing of the 

ACTIFLO system. The pilot plant consists of three mixing tanks with a total volume of 12.5 

m and two parallel hopper settling tanks each having a surface area of 1.2 m2

. Each 

mixing tank is equipped with an axial flow mixer designed to allow for optimal coagulation 

and flocculation. A rubber lined centrifugal pump extracts the sand/sludge slurry from the 

bottom of the hopper settling tanks. The major operating parameters of the pilot plant at 

the nominal flow rate of 150 m3

/h are summarized in Table 5.1.b-6. 

Table 5.1.b-6. Major-operating parameters of ACTIFLO pilot plant at nominal flow rate of 

150 m3

/h 

Parameters Values Units 

Coagulation Tank Detention Time 1 min 

Injection Tank Detention Time 1 min 

Maturation Tank Detention Time 1 min 

Settler Overflow Rate 150 m/h 

Microsand Recirculation Rate 8 % 

Microsand Concentration 3—6 kg/m3 

Velocity Gradient (G) in Coagulation Tank 200—300 1/s 

Velocity Gradient (G) in Injection Tank 200—300 1/s 

Velocity Gradient (G) in Maturation Tank 150—200 1/s 

47
 
Case Studies 

Primary Wastewater Treatment Application - Wastewater Reuse for Agricultural Irrigation 

in Mexico City, Mexico. Wastewater reuse for agricultural irrigation must be treated for the 

removal of Helminth eggs and protozoa cysts, and elimination of pathogenic bacteria and 

inactivation of enteric viruses. Of great importance is the removal of Helminth eggs. They 

are extremely persistent, surviving in harsh environmental conditions, and due to their 

latency period, are transmitted primarily through uncooked food. The Helminth disease 

transmission has been identified as the top-priority heath problem of the wastewater reuse 

in developing country. The World Heath Organization (WHO) has recommended that the 

Helminth egg concentration be under 1 Helminth egg per liter for all kinds of agriculture 

(WHO, 1989). 

After several studies, the Mexican authorities had determined that the best and most 

economical way to reuse the Mexico City wastewater for agriculture irrigation was to use 

advanced primary treatment. Due to the existing situation and the extremely large flow of 

74.5 m3

/s (1,700 MGD), the Mexican authorities decided to evaluate new treatment 

technologies and organized pilot studies to see their performances. A six-month pilot test 

was conducted to investigate the performance of the ACTIFLO process in 1997 (Le Poder 

1997). 

The results are summarized in Table 5.1.b-7 and in Figure 5.1.b-4 and 5.1.b-5 and show 

that it was possible to obtain very low SS (25 mg/l in average) and Helminth eggs residual 

concentration (1 egg per test in average) even at very high overflow (up to 200 m/h) and 

low chemical dosage (<60 ppm of alum). It confirms that the ACTIFLO process, even 

without the downstream filtration process, is able to produce an effluent water with less than 

1 egg/liter in daily average samples the majority of the time, and less than 5 eggs/liter in all 

cases, which is the maximum amount permitted for certain kinds of agriculture. 

Table 5.1.b-7. Removal efficiency from 20 daily samples in primary treatment in Mexico 

City, Mexico 

Helminth Fecal 

SS
 Eggs Coliform COD Total N Total P Sulfurs 

mg/l Eggs/l /100 ml mg/l mg/l mg/l mg/l 

Influent 302 24.3 8x108 442 18.4 10.3 7.1 

Effluent 24.7 0.99 1.4x108 172 15.5 1.9 1.3 

Removal (%) 91.7 96 82.8 61.1 15.8 81.6 81.7 

48
 
0.1 

1 

10 

100 

Date 

Helminths Eggs (Eggs/liter) 

0 

50 

100 

150 

200 

250 

Rise Rate (m/h) 

Rise Rate Influent TSS Effluent TSS 

04/10/97 04/16/97 05/14/97 05/21/97 05/28/97 06/04/97 06/11/97 06/24/97 06/25/97 06/26/97 07/02/97 07/03/97 07/14/97 

Figure 5.1.b-4. Helminth eggs removal in primary wastewater treatment in Mexico City, 

Mexico 

0 

100 

200 

300 

400 

500 

600 

700 

6/4/97 

7/14/97 

Date 

TSS (mg/l) 

0 

50 

100 

150 

200 

250 

Rise Rate (m/hr) 

Rise Rate 

Influent TSS 

Effluent TSS 

4/10/97 4/16/97 5/14/97 5/21/97 5/28/97 6/11/97 6/24/97 6/25/97 6/26/97 

7/2/97 7/3/97 

Figure 5.1.b-5. SS removal in primary wastewater treatment in Mexico City, Mexico 

CSO Treatment Application in Galveston, Texas, Pilot Study. A pilot study was 

conducted by CDM (1998) in Galveston, TX to evaluate the effectiveness of the ACTIFLO 

process under simulated CSO conditions. In order to simulate CSO conditions raw 

wastewater was pumped through a screen into a holding basin where it was subsequently 

blended with the secondary effluent from the existing WWTF prior to entering the pilot unit. 

The effect of coagulant dosage on SS, COD and BOD5 removal efficiencies is shown in 

Table 5.1.b-8. The pilot unit was operated at an overflow of 75 m/h, a polymer dosage of 

1.0 mg/l and a pH of coagulation of 6.0 while the coagulant dosage (FeCl3) was varied 

from 75—125 mg/l. As shown in the figure, the SS and COD removal efficiencies 

improved as the FeCl3 was increased from 75 mg/l to 100 mg/l. The removal efficiencies 

of SS and BOD5 remained constant as the FeCl3 was increased from 100 mg/l to 125 

mg/l. Therefore, the optimum coagulant dosage was approximately 100 mg/l. 

49
 
Table 5.1.b-8. Effect of Coagulant Dosage on Removal Efficiencies in Galveston, TX 

FeCl3 SS COD BOD5 

Dose Raw Settled % Raw Settled % Raw Settled % 

Mg/l Mg/l Mg/l Rem. Mg/l Mg/l Rem. Mg/l Mg/l Rem. 

75 123 16.9 86% 500 112 78% 105 39 63% 

100 147 13.4 91% 600 105 92% 123 35 72% 

125 135 10.9 92% 300 105 65% 195 52 73% 

In the event of CSO conditions a process that is primarily used during overflow periods will 

have to start up and achieve steady state very quickly. Figure 5.1.b-6 illustrates the rapid 

start up capabilities of ACTIFLO process. The first SS and COD samples were taken after 

two hydraulic retention times (18 min), the time needed to achieve steady state. After the 

process reached steady state the SS and COD removal efficiencies were 88% and 75% 

respectively, and the average SS and COD removal efficiencies over the testing period 

were 91% and 73%, respectively. The short overall retention time of the process allows the 

rapid start up capabilities of ACTIFLO process. During this experiment the pilot unit was 

operated at an overflow rate of 75 m/h, a FeCl3 dosage of 100 mg/l and a polymer dosage 

of 1.0 mg/l. 

50 

60 

70 

80 

90 

100 

15:00 15:10 15:20 15:30 15:40 15:50 16:00 16:10 16:20 16:30 

Time 

Average Percent Removal 

TSS 

COD 

The process was started up at 15:00 

Figure 5.1.b-6. Start-Up Efficiency of ACTIFLO Process 

CSO Treatment Application in Cincinnati, Ohio Pilot Study. Kruger Inc conducted a pilot 

study using the ACITFLO®

 process for the Metropolitan Sewer District at the Mill Creek 

WWTF in Cincinnati, OH (Kruger 1998). The testing period was from April 14, 1998 

through April 29, 1998. This pilot study was conducted to demonstrate the effect of 

overflow rate on the percent removal of turbidity, BOD5, and total phosphorous during CSO 

conditions. On April 17, 1998 the pilot unit was operated at a polymer dosage of 

50
 
1.0 mg/l, a FeCl3 dosage of 35 mg/l while the overflow was varied from 75 m/h to 150 m/h. 

Table 5.1.b-9 depicts the effect of overflow rate on the removal efficiencies. 

Table 5.1.b-9. Effect of overflow rate on removal efficiencies, Cincinnati, OH 

Rise Turbidity BOD5 Total Phosphorous 

Rate Raw Settled % Raw Settled % Raw Settled % 

m/h NTU NTU Rem. mg/l mg/l Rem. mg/l mg/l Rem. 

75 64 2.06 96.6% 125 23 81.6% 0.60 ND1 >99% 

100 72 2.35 96.8% 139 20 85.6% 1.03 ND >99% 

150 80 4.65 94.2% 71 20 71.8% 2.99 ND >99% 

1 ND = None Detected 

ACTIFLO process achieved greater than 94% removal of turbidity and > 70% removal of 

BOD5 and >99% removal of total phosphorous. The process performance was not 

compromised at the higher overflow rates in regards to total phosphorous removal. By 

comparing the settled water results, higher overflow rates do not adversely effect the 

process performance. 

An overnight run was performed to demonstrate the stability and the performance of the 

process during CSO conditions. The results of this overnight run are shown in Figure 

5.1.b-7. During the night of April 22 and the morning of April 23, the process was operated 

at an overflow rate of 100 m/h, a polymer dosage of 1.0 mg/l to 1.3 mg/l and a FeCl3 

dosage of 45 mg/l to 100 mg/l. The SS removal varied from 60% to 97.2% with an 

average percent removal of 78.8% and a standard deviation of 7%. The BOD5 percent 

removal varied from 17.1% to 71.6% with an average percent removal of 45.4% and a 

standard deviation of 18.2%. 

0 

20 

40 

60 

80 

100 

120 

0 5 10 15 20 25 Hour 

Percent Removal 

TSS BOD5 

Figure 5.1.b-7. Twenty-four Hour Continuous Demonstration Run, Cincinnati, Ohio 

51
 
Conclusions 

As a primary clarification process, the ACTIFLO process can achieve high removal 

efficiencies of pollutants including SS and Total Phosphorous. Additionally, the quick startup and short detention time of ACTIFLO process make it a viable alternative for CSO 

treatment. 

This desktop analysis demonstrates that while the ACTIFLO process may be a viable 

treatment process, the capital and O&M costs are high versus conventional alternatives. 

One of the primary benefits of the ACTIFLO system is the extremely small footprint of the 

process. This can make the ACTIFLO system a more cost-effective alternative for sites 

where space for large storage facilities is limited. 

This process can essentially be adapted and retrofitted into existing primary treatment 

facilities and for CSO control. 

52
 
5.2 Retrofitting Existing WWF Storage Tanks to Provide Enhanced Settling/ 

Treatment and Post-Storm Solids Removal. Spring Creek AWPCP, New York, NY 

Background 

The Spring Creek Auxiliary Water Pollution Control Plant (AWPCP) is a wet-weather 

combined sewage detention facility with an existing maximum storage volume of 

approximately 49,200 m3 (13 MG): 37,900 m3 (10 MG) in basin storage and 11,400 m3 

(3 MG) in influent barrel storage. The AWPCP facility is located on Spring Creek in 

Queens, New York, approximately 1 mile east of the 26 Ward Water Pollution Control 

Plant (WPCP), near the Brooklyn-Queens border. The function of the AWPCP is to 

capture combined sewage, above elevation -7.00, from tributary drainage areas in 

Brooklyn and Queens. Flow is conveyed to the plant by the four overflow barrels from 

the Autumn Avenue regulator located in Brooklyn, and by two overflow barrels from the 

157th Avenue regulator located in Queens. The plant has six basins that fill with 

combined sewage to a depth of 4.6 m (15 ft) and up to a maximum elevation of +1.00, 

depending on Jamaica Bay tide elevations. Once the basins fill to elevations of -0.50 or 

greater depending on tides, CSO is released into the bay through the flap gates at the 

south end of the basins. As flow recedes in the sewers, combined sewage retained 

within the basins and influent barrels above elevation -7.00 flow by gravity back into the 

sewer system for treatment at the 26 Ward WPCP. Retained combined sewage below 

elevation -7.00 is screened and pumped back into the sewer system through the 0.6 m 

(24 in.) diameter plant effluent sewer. The layout of AWPCP is shown in Figure 5.2-1. 

Overview 

This desktop analysis evaluates the concept of retrofitting the AWPCP to provide 

enhanced treatment of stored WWF and removal of solids accumulated in the basin 

following storm events. WWF is presently retained in the AWPCP to minimize wetweather overflow through equalization, however, no provision for treatment is provided. 

The residual solids and debris deposited within the storage basins are currently cleaned 

by a system of traveling bridges equipped with a horizontal spray header and a water 

cannon. Each bridge, one per basin, is operated in multiple passes over the basin 

length and uses spray water (brackish bay water) to clean the basin walls and floor. 

Two alternatives, cross-flow plate settlers and chemical addition, were evaluated to 

determine the dual effectiveness of retrofitting AWPCP to serve as an equalization 

system and to enhance the removal of solids prior to discharge to Jamaica Bay. 

The criteria for retrofitting the AWPCP for enhanced treatment were selected to allow 

conceptual design of each retrofit. The primary design factor of these alternatives is the 

peak flow rate that must be conveyed and treated. The one-year frequency, two-hour 

duration design storm was selected for this analysis. The peak flow rate associated 

with the one-year design storm is 66 m3

/d (2,330 cfs) based on SWMM modeling of the 

combined sewer system tributary to Spring Creek and the 26 Ward WPCP. 

53
 
5.2-1 Spring Creek AWPCP Layout 

54
 
The existing dimensions of the AWPCP and the design flow rate for the proposed 

treatment facilities are shown in Table 5.2-1. 

Table 5.2-1. Existing Dimensions of the AWPCP and Design Flow Rate 

Design Data Quantity 

Design Storm Flow (one-year storm)
 63.2 m3/s (2,230 cfs)
 

Number of Retention Basins
 6
 

Flow / Basin
 11.0 m/s (388 cfs)
 

Length of Basin
 152.4 m (500 ft)
 

Width of Basin
 15.2 m (50 ft)
 

Depth of Basin
 3.7 m (12.2 ft)
 

Basin Volume
 8,670 m3 (306,145 ft3

)
 

Total Volume
 52,000 m3

 (1,836,158 ft3)
 

Basin Surface Area
 2,322 m2

 (25,000 ft2)
 

Total Surface Area
 13,935 m2 (150,000 ft2)
 

Cross Sectional Area / Basin
 56.7 m2

 (610 ft2)
 

Surface Loading Rate
 410 m3/d/m2 (10,067 gpd/ft2

)
 

Hydraulic Detention Time
 15 min
 

5.2.a. Cross-Flow Plate Settlers 

Cross-flow plate settlers use lamellar clarification to remove solids from wastewater. 

This process takes advantage of the decantation principle for SS that are denser than 

water for clarification. Cross-flow plate settlers have performed as well as conventional 

clarifiers, with a smaller footprint to achieve a similar level of settling. 

Cross-flow lamellar clarification are principally used for the removal of oils and grease 

present in residual liquids emanating from industrial activities in the petrochemical, 

chemical, mechanical, metallurgical, paper-processing and food-processing sectors. 

The cross-flow plate settler technology has also been successfully utilized in 

conventional primary clarifiers and for gravity settling of surface water runoff. 

Case History: Shell Montreal Canada 

The Shell Company in Montreal, Canada retrofitted their existing storage basin with 

cross-flow plate settlers to improve the water quality of the effluent prior to discharge. 

This concrete storage basin collects surface runoff from the refinery area. The 

55
 
2

parameters that were used for design of the cross-flow plate settlers are shown in Table 

5.2.a-1. 

Table 5.2.a-1. Design Parameters for the Cross-Flow Plate Separator 

Parameter Quantity 

Treatment Capacity (peak flow)
 7,570 L/min (2,000 gpm)
 

Influent TSS Concentration
 200 mg/L
 

Average TSS Density
 1,159 mg/L
 

Temperature
 20 deg. C
 

pH
 8 to 9
 

Overflow Rate
 404 m3

/d/m2 (9,930 gpd/ft2)
 

The concrete tanks were modified to include cross-flow plate settlers that provided a 27 

m (290 ft2) surface area for a total projected surface of 223 m2 (2,400 ft2). The 

installation included: 

• Inlet deflectors 

• Inlet screening 

• Two cross-flow plate cells 

• A V-notch weir 

• An adjustable skimmer 

• An effluent weir 

The layout of the retrofitted settling tanks is shown on Figure 5.2.a-1. 

Limited evaluation of this retrofit has shown that SS concentrations have been reduced 

to between 100—125 mg/L throughout a range of influent flow rates and water qualities. 

This retrofit would improve the treatment capacity of the existing settling tank without 

the requirement for costly expansion or the construction of new settling tanks to achieve 

comparable levels of SS removal. Cost information for this retrofit is unavailable. 

Proposed Retrofit of AWPCP with Cross Flow Settlers 

This retrofit consists of converting the existing Spring Creek basins to include cross-flow 

lamellar plate settlers to provide treatment through gravity settling of dense particles. 

The use of inclined plates significantly increases the ratio of settling surface to footprint 

that would be required for conventional primary clarifiers. The capacity of primary 

clarifiers is primarily a function of the surface area rather than depth. The settling area 

is a function of the surface area. Theoretically, the settling area of a given spatial area 

could be increased by stacking horizontal plates to achieve a settling area equal to the 

surface area times the number of plates. However, this arrangement does not allow for 

removal of the accumulated sludge. 

56
 
5.2.a-1 Layout of Shell Co. Cross-Flow Plate Settler, Montreal Canada 

57
 
Inclining the plates would make sludge removal possible. Each plate would have an 

effective settling area equal to its horizontal projection. The projections of inclined 

plates packed close together (e.g. 4 in. apart) overlap. The total settling area is 

determined by adding the horizontal projections. This can increase the theoretical 

settling area by 10 times the settling area compared to a conventional clarifier (Huebner 

1979). 

There are two major force vectors acting on SS contained within wastewater in the 

cross-flow separator. The force of gravity acts to pull the SS down while the velocity 

vector pushes the SS up along the length of the plates. The resultant of these forces 

directs the SS to the plate where particles impinges on the plate and come in contact 

with other particles sliding downward along the plate. The flow between the plates is 

maintained in the laminar range to promote settling. 

The treatment process begins with the screened influent entering the lamellar zone that 

consists of cross-flow cells. Particle settling occurs by gravity according to Stoke's Law 

that describes the velocity of a particle falling through a fluid. Stoke’s Law states that 

the velocity of a particle is equal to the square of the particle radius. 

In the design of sedimentation basins, the usual procedure is to select a particle with a 

terminal velocity Vc and to design the basin so particles that have a terminal velocity 

equal to or greater than Vc will be removed. 

The lamellar cell design is a practical application of Hazen's theory on settling that 

provides a large equivalent surface of separation (as noted by the projected horizontal 

area for each plate). Hazen’s theory is stated by Huebner (1979) where the velocity is: 

V = average linear velocity = 

q (2) d � w 

q = volumetric flow rate 

d = distance between plates 

w = width of plates 

Q = angle of elevation to the horizontal 

Vs = particle settling velocity 

ds = maximum settling distance 

L = length of plates 

Particles are removed from the flow stream when its settling time is less than its flow 

through time. 

t £ t (3) s q

d 

t = s (4) s Vs 

58 
L L L � d � w 

t = = = (5) q V q d � w q 

The maximum settling distance, ds can be calculated as a function of the distance 

between the plates, d and the angle of inclination of the plate, Q. 

d d = (6) s Cos Q 

Which equates to: 

d Cos Q 

£ L � d � w (7) 

V q s 

and further simplifies to: 

q £ Vs Cos Q (8) L � w 

Hazen’s theory provides the surface loading rate (Y) for a conventional clarifier, as: 

Y = 

q (9) 

L � w 

which is equivalent to the simplified inclined plate formula derived above in that q is the 

total flow divided by the number of plates. The conventional surface loading rate occurs 

when the angle Q goes to zero, resulting in a Cos Q of 1. 

The principal design parameters of cross-flow plate settlers are: 

Wastewater flow rate - Settling systems are designed to accommodate the peak flow 

rate, but the maximum removal efficiency is normally achieved at a lower design flow 

rate. 

Settling rate of solids – The settling rate of solids is measured in laboratory tests. A 

design velocity is then selected according to the required removal efficiency. The 

required settling area is calculated by dividing the design flow rate by the design 

velocity. 

Angle/spacing of the plate – The angle of the plates typically varies from 55—65 

degrees and the spacing varies from 2.5—10.2 cm (1—4 in.) depending upon the 

nature and the concentration of solids and space limitations. The 10.2 cm (4 in.) 

spacing is commonly used in wastewater applications. 

A depiction of the cross-flow plate is shown on Figure 5.2.a-2. 

59
 
Figure 5.2.a-2. Horizontal Projections of Inclined Cross-Flow Lamellar Plates 

Cross Flow Plate Settler Equipment 

The cross-flow plate settlers selected for this retrofit are from Aquarius Services & 

Technologies Inc., Quebec. The major equipment of the cross-flow lamellar plate 

settlers is shown in Figures 5.2.a-3 and 5.2.a-4, and include the following: 

1. Flow Deflectors – Flow deflectors at the inlet of each basin dissipate the energy 

generated by the influent flow. This encourages flocculation and distributes the flow 

evenly along the width of the basin and wet perimeter of the lamellar cell. 

2. Inlet Bar Screen - Floating material that may clog the lamellar cells is removed by a 

inlet bar screen with 5.1 cm (2 in.) square openings 

3. and 4. Cross Flow Cells - Cross-flow type lamellar cells consist of inclined plates at a 

60� angle and are constructed of either aluminum, stainless steel or synthetic materials. 

Spacing between plates may vary according to the design requirements. For the 

purpose of this retrofit, stainless steel plates with 10.2 cm (4 in.) spacing were selected 

based on the manufacturer’s suggestions for this installation. 

5. Skimming Device - Oils and hydrocarbons are skimmed from the surface through an 

adjustable skimmer. 

6. Scum Baffle - A scum baffle is installed in back of the skimmer to retain the floatables 

for collection and disposal through the skimmer. 

7. Weir - A V-notch weir is installed at the discharge end of the basin to regulate the 

liquid level in the basin. 

8. Sludge Vacuum Collector - Each sludge silo is equipped with a sludge vacuum 

collector to transfer and dispose of collected sludge. 

60
 
Figures 5.2.a-3. Plan View of Aquarius Cross-Flow Plate Settler 

61
 
Figure 5.2.a-4. Profile of Aquarius Cross-Flow Plate Settler 

62
 
Following the completion of the storm, the basins would be automatically washed and 

the sludge will be transferred for disposal. 

The hydraulic design criteria for the cross-flow plate settler technology is shown on 

Table 5.2.a-2. 

Table 5.2.a-2. Cross-Flow Plate Settler Design Data 

Design Data Quantity 

Hydraulic Loading Rate* 0.15 m3/s/100 m2 (0.5 cfs/100 sq. ft) 

Overflow Rate 132 m3/day/m2 (3,250 gpd/sq. ft.) 

* Loading Rate furnished by Aquarius Services & Technologies Inc. Quebec 

Under the above loading rates and available basin surface area, the one-year design 

flow rate of 63 m3/s (2,230 cfs) would not be accommodated by a cross flow plate settler 

system. The proposed system is capable of treating a maximum flow rate of 1,835,725 

m3

/d (485 MGD) or 32% of the design flow rate based on the physical size of the Spring 

Creek basins and the limited hydraulic loading rates of the cross-flow plate settlers. The 

proposed facilities would provide treatment under a small percentage of wet-weather 

events observed at Spring Creek and would create a hydraulic limitation to WWF 

entering the facility. Therefore, retrofit with cross-flow plate settlers is not considered 

viable for this site. 

5.2.b. Chemical Addition to AWPCP 

This desktop analysis involves retrofitting the AWPCP by chemical addition to provide 

enhanced settling/treatment. Chemical addition describes a method of treatment where 

chemicals are added to enhance settling of particles present in the wastewater. Treated 

water is then decanted and conveyed for further treatment or discharged. The resultant 

sludge is collected for disposal and can be de-watered to reduce the volume of solids. 

Chemical precipitation can be used to remove metals, fats, oils and greases (FOG), SS 

and some organics. It can also to be used to remove phosphorus, fluoride, ferrocyanide 

and other inorganics. 

Conventional wastewater clarification processes primarily involve the destabilization and 

subsequent removal of colloidal SS materials that are not readily removed by gravitation 

alone. These suspended materials can be natural or synthetic organic or inorganic 

compounds, microorganisms, and/or viruses that typically range in size from 10-4 �m to 

1,000 �m. In most natural systems, the stability of colloidal suspended materials is 

attributed to a net negative surface charge that causes individual particles to repel each 

other and remain in suspension. To counteract these repulsive forces a highly charged 

ionic chemical, such as alum (Al2 (SO4)3), FeCl3, ferric sulfate (Fe2(SO4)3), polyaluminum chloride (PACl), is added to bring about a net reduction in the repulsive force 

between the suspended materials. This process, termed coagulation, results in the 

63
 
destabilization and/or attraction of the SS to form chemical floc. Although destabilized, 

the chemical floc may remain in solution due to its extremely small size and low mass. 

These floc, often called “pin floc,” are difficult to remove from the treated water without 

first aggregating the smaller particles together into larger, heavier, more settleable floc 

by flocculation. 

Flocculation is the term used to describe the aggregation of pin floc into larger, heavier, 

more settleable floc. This is most readily accomplished with the addition of a chemical 

(flocculant aid) polymer. Such polymers are available from many sources in both liquid 

and dry form with varying molecular weights and net surface charges. In most cases 

selection of the best polymer for a specific application and set of conditions is largely a 

matter of trial and error. Larger, more settleable flocs form by inter-particle 

polyelectrolyte bridges between floc and destabilized solids. Although the formation of 

inter-particle bridges can take many forms, the end result is the enmeshment of large 

volumes of particles to form large dense floc that are more readily removed by 

gravitational settling. 

The use of chemical addition would enhance the settling of discrete particles and 

suspended matter thereby increasing the overall removal of BOD5 and SS. This would 

be accomplished by retrofitting the existing basins with chemical feed, mixing and 

contact facilities. The retrofit would also include dedicated sludge collection and 

removal facilities. Chemical coagulation of raw wastewater before sedimentation 

promotes flocculation of finely divided solids into more readily settleable flocs, thereby 

increasing SS, BOD5, and phosphorous removal efficiencies. Sedimentation with 

coagulation may remove 60—90% of the SS, 40—70% of the BOD5, 70—90% of the 

phosphorous, and 80—90% of the bacterial loadings. In comparison, sedimentation 

without coagulation may remove only 40—70% of the SS, 25—40% of the BOD5, 5— 

10% of the phosphorous loadings and 50—60% of the bacteria loading. 

Chemical Coagulants 

Advantages of coagulation include greater removal efficiencies, the ability to treat higher 

flow rates, and more consistent performance. Disadvantages of coagulation include an 

increased mass of primary sludge that is often difficult to thicken and de-water, and 

increased in operational cost and operator attention (WEF 1992). Iron salts, particularly 

FeCl3, are the most common coagulant used for primary treatment. WWTF using lime 

for primary treatment have found that it produces more sludge than do metal salts and 

is more difficult to store, handle and introduce into the waste stream. Coagulants 

should be selected for performance, reliability and cost and evaluations made to 

determine dosages and effectiveness. 

64
 
Case Histories 

Table 5.2.b-1 lists coagulant dosages and performance data for six (6) treatment plants 

that employ chemical coagulation for advanced primary performance in sedimentation 

tanks. The Joint Water Pollution Control Plant (JWPCP) at Carson, California evaluated 

30 different polymers and found anionic polymers to be the most effective type for 

enhancing primary sedimentation. The Hyperion plant in Los Angeles, California uses 

to great success FeCl3 and an anionic polymer at the headworks before primary 

sedimentation. This process increased primary sludge production 45% through 

improved SS and colloidal matter removal. This increase was partially offset by a 

decrease in waste activated sludge production. 

The Los Angeles and San Diego sedimentation tanks with chemical coagulation operate 

with overflow rates between 69.3—81.5 m3

/d/m2 (l700—2000 gpd/ft2) as compared to 

conventional overflow rates of 32.6 m3/d/m2 (800 gpd/ft2) for primary settling. Research 

in Sarnia and Windsor, Ontario indicated that overflow rates up to 97.8 m3

/d/m2 (2,400 

gpd/ft2) did not significantly affect effluent quality. 

Design Considerations 

The design flow rate for this retrofit is 66 m3/s (2,330 cfs) as shown in Table 5.2-1 with a 

corresponding overflow rate of 410.1 m3/d/m2 (10,067 gpd/ft2) based on the dimensions 

of AWPCP. This high surface SOR is greater than what would typically be applied for a 

chemical coagulation system; however, for the purposes of this retrofit it is assumed 

that during the higher SOR’s, a reduced efficiency will result. It is important to note that 

the frequency at which these high SOR’s are likely to occur are expected to be relatively 

infrequent (e.g., once per year). 

The proposed retrofit consists of equipping the existing influent barrels with a chemical 

feed and mixing system. These facilities would provide the necessary mixing and 

contact for the rapid mix and flocculation steps, which include the addition of FeCl3 and 

an anionic polymer. An important element in this analysis is an evaluation of the 

hydraulic and physical parameters dictated by the existing basins as depicted on Table 

5.2-1. 

This retrofit would include the specific elements outlined below. Based on a ratio of 452 

kW / m3

/d (0.16 HP/MGD) at a peak flow of 953,820 m3/d (252 MGD) per barrel, a total 

of 30 kW (40 HP) is necessary to achieve the mixing required for the FeCl3 per barrel. 

There will also be 2 mixers per basin for mixing the polymer. Therefore, 24—15 kW (20 

HP) mixers (4 /barrel) are necessary for the 6 basins. Chemical storage requirements 

are based on the need to treat approximately 2,500,000 m3

 (660 MG) of CSO per year. 

Storage of FeCl3 (37% solution) and polymer (0.25 mg/l) to treat a month (on average) 

of CSO would require approximately 16 m3 (4,200 gal) of chemical storage. 

65
 
Table 5.2.b-1. Coagulant Dosages and Performance Data for Six Treatment Plants That Employ Chemical Coagulation 

for Advanced Primary Performance in Sedimentation Tanks 

Advanced Primary Performance 

BOD TSS Chemical Addition 

Flow Influent Effluent Rem. Influent Effluent Rem. Type Conc. Duration 

Location (m3 /day) (MGD) mg/L mg/L % mg/L mg/L % mg/L 

Point Loma FeCl3 35
 

City of San Diego 722,935 191 276 119 57 305 60 80.3 Anionic Poly 0.26 Continuous
 

Orange County FeCl3 20 8 hours
 

Plant No. 1 227,100 60 263 162 38 229 81 64.6 Anionic Poly 0.25 Peak Flow
 

Orange County FeCl3 30 12 hours
 

Plant No. 2 696,440 184 248 134 46 232 71 69.4 Anionic Poly 0.14 Peak Flow
 

JWPCP
 

Los Angeles County 1,438,300 380 365 210 42 475 105 77.9 Anionic Poly 0.15 Continuous
 

Hyperion FeCl3 20
 

Los Angeles City 1,400,450 370 300 145 52 270 45 83.3 Anionic Poly 0.25 Continuous
 

Sarnia FeCl3 17
 

Ontario, Canada 37,850 10 98 49 50 124 25 79.8 Anionic Poly 0.30 Continuous
 

66
 
Chemical feed pumps will be necessary to deliver the correct dosing of chemical during 

wet-weather events. The FeCl3 solution will require a feed rate of approximately 0.05 

m3/min (13 gpm) per barrel at peak flow; resulting in 2 metering pumps per barrel for a 

total of 24. The pumps would be hose (or peristaltic type) pumps. Half as many feed 

pumps would be required to deliver the polymer to the influent barrels. 

Operators could control the chemical feed system using a new control panel and 

SCADA system housed in the existing control room. The capital costs for this retrofit 

are shown on Table 5.2.b-2. 

Table 5.2.b-2. Estimated Capital Costs Chemical Addition Retrofit 

Description Cost 

General Conditions $25,000 

Sitework $350,000 

Chemical Storage $25,000 

Chemical Feed System $75,000 

Chemical Mixers (4 in each of 6 basins) $720,000 

Instrumentation and SCADA System $25,000 

Total Construction Cost $1,220,000 

Project Services1 $240,000 

Total Project Cost $1,460,000 

1

 Includes Design & Construction Engineering, Legal, and Fiscal Expenses 

Does not include contingencies and right-of-way costs 

Operational Requirements 

The existing AWPCP operates intermittently based on wet-weather events. Following 

CSO events, the facility is de-watered by gravity and pumping. These operations would 

not change under the proposed retrofit. The following list summarizes the areas of 

expected O&M cost increases at the facility: 

•	 Increased labor to perform wet-weather process adjustments and clean-up following 

a wet-weather event. 

• Power for chemical feed and mixing systems , and 

• Chemicals 

The costs are based on the facility treating approximately 2,498,100 m3 (660 MG) of 

CSO per year and maintaining a chemical storage of approximately one month. The 

annual volume of CSO is based on modeling performed for a typical year using the 

Storm Water Management Model (SWMM). Hourly rainfall data were used from a local 

NWS Weather Station as input to the model. It is estimated that approximately 100 wetweather events occur per year and that each event is approximately 6 h in duration. 

The O&M costs are shown in Table 5.2.b-3. 

67
 
Table 5.2.b-3. Estimated Chemical Addition O&M Costs 

Description Estimated Annual Cost 

Coagulant Consumption 

Coagulant Type Ferric Chloride 

Coagulant Dosage (ppm) 20 

Design Flow Rate (MGD) 1,510 

Coagulant Consumption (lbs./day) at Peak Flow 252,000 

Coagulant Consumption (lbs/month) 9,200 

Coagulant Consumption (tons/year) 44 

Coagulant Cost ($/ton) 200 

Coagulant Cost ($/year) $8,800 

Polymer Consumption 

Est. Polymer Dosage (ppm) 0.25 

Design Flow Rate (MGD) at peak flow 1,510 

Polymer Consumption (lbs./day) at Peak Flow 3,100 

Polymer Consumption (lbs./month) 110 

Polymer Consumption (tons/year) 0.53 

Polymer Cost ($/ton) 3,500 

Polymer Cost @ 100% /24/365 ($/year) $1,848 

Labor 

Cleanup/Operations $21,000 

Energy Consumption 

Rapid Mixer HP per Basin 30 

Chemical Feed HP per Basin 2 

Total HP (for 6 basin compartments) 192 

kWh/year 115,200 

Total Annual Energy Cost ($/year) $6,912 

Total Annual Operation Cost ($/year) $38,600 

6 hours per event 

Conclusions 

This retrofit utilizes the existing storage facilities to provide chemically enhanced 

sedimentation of WWF. This could result in the removal of 60—90% of the SS, 40— 

70% of the BOD5, and 30—60% of the COD from the effluent to the receiving water. 

This is an expected 58% increase over current performance without chemical addition. 

At the one-year design flow, the facility will exceed typical SORs by a factor of 4—5, 

however, these would occur relatively infrequently. 

68
 
5.2.c. Retrofitting Existing WWF Storage Tanks For Post-Storm Solids Removal, 

Spring Creek AWPCP, New York, NY 

In 1997, CDM conducted a basin-cleaning prototype testing program at the Spring 

Creek AWPCP in New York City (CDM 1997). The prototype testing was performed in 

response to the New York City Department of Environmental Protection's (NYCDEP) 

desire to investigate alternatives to the existing traveling bridge-cleaning system. This 

analysis was performed to evaluate the results of the prototype testing and provide 

recommendations for basin cleaning system design. 

Background 

The residual solids and debris deposited within the storage basins are actively cleaned. 

Currently, for cleaning of solids and debris within the basins, the facility uses a system 

of traveling bridges equipped with a horizontal spray header and a water cannon. Each 

bridge, one per basin, is operated in multiple passes over the basin length and uses 

spray water (brackish bay water) to clean the basin walls and floor. 

The basin cleaning alternatives evaluated through prototype installation and testing at 

the facility were: 

• Tipping buckets 

• Fixed spray nozzles 

• Orifice headers 

• Flushing gate (Hydroself Fluid Flap) 

• Re-suspension/mixing (used to enhance any of the above) 

Objectives and Scope 

The effectiveness of the cleaning systems at the Spring Creek AWPCP were evaluated 

to establish design criteria for feasible alternatives for full implementation and to 

develop operational requirements and operational and capital costs associated with 

each of the systems. The data collected from the prototype study was used to either 

determine or indicate the following: 

• Effectiveness of floor cleaning 

• Effectiveness of wall cleaning 

• Water usage 

• Power usage 

• Overall cleaning time required 

• Special operating requirements 

• Safety concerns (if any) associated with each system 

• Cost 

Each of the systems listed above was installed in test areas within three of the six 

basins at Spring Creek. As discussed below, each system was arranged in different 

69
 
configurations and operated during actual basin cleaning events for a total of 16 

cleaning test events, over the period from July 1 to December 16, 1997. The cleaning 

systems were tested under conditions where solids accumulation varied widely. Each 

system was operated under a variety of cleaning water volumes, flow rates, and 

pressures. As the testing proceeded, it became apparent which technologies were not 

providing effective cleaning. As a result, the original test protocol was modified to focus 

on, and optimize the performance of the most effective systems. 

The actual test conditions and the equipment used in the study are summarized below. 

Basin Cleaning Prototype Equipment and Operation. The prototype testing consisted of 

16 test runs, from July 1, 1997 through December 17, 1997. The prototype fieldwork 

was performed in conjunction with NYCDEP's routine basin cleaning operations to 

benchmark the existing cleaning operations at the facility. 

The following parameters were monitored and recorded during the prototype testing: 

• Solids depth within basins 

• Equipment test parameters (flow rate, total wash-water volume, pressure, etc.) 

• System cleaning effectiveness 

• Cleaning duration 

The overall cleaning effectiveness was based on a qualitative assessment of the degree 

of cleaning and the approximate percentage of the floor area cleaned. The test area 

was judged to be clean when approximately 99% of the test area was without solids. 

Water consumption and operating pressures were measured using flow meters and 

pressure gauges set up throughout the system. Cleaning time was measured for the 

prototype systems and then estimated regarding full-scale implementation of each 

system. 

An evaluation of the existing basin cleaning operation was performed and provided a 

benchmark for the prototype cleaning systems. This benchmarking was performed by 

observing the cleaning operations, documenting the cleaning time and using the rated 

pump capacity to calculate the flow rate and water used by the existing traveling 

bridges. 

Using the traveling bridges, the basins were generally cleaned one side at a time. 

Typically, the traveling bridge requires a minimum of two passes per side. The first 

pass loosens the solids and provides some cleaning, while the latter passes provide for 

a more thorough cleaning. A manually controlled water cannon breaks up large heavy 

solids. In addition to cleaning the basin floor, the water cannon cleans the sidewalls of 

the basin and the walkways of the basin area. 

70
 
General Prototype Systems Installation and Operation. Five prototype cleaning 

systems were tested: tipping buckets, flushing gate, fixed spray headers with nozzles, 

fixed spray headers with orifices, and a submersible mixer. The prototype systems 

were located in basins 1, 2 and 3. 

Bay water was used as the source of cleaning water for the prototype systems. 

Cleaning water was supplied from a single, 37.3 kW (50 HP), three stage vertical 

turbine wash-water pump, installed in the existing spray water channel between basins 

1 and 2. The pump was sized for a flow rate of 3 m3/min (800 gpm) against a head of 

50.3 m (165 ft). A PVC piping system distributed wash-water to the various cleaning 

systems. A flow schematic of the wash-water equipment and piping is provided in 

Figure 5.2.c-1. Flow meters, pressure gages, and butterfly valves were installed 

throughout the cleaning-water piping-system to provide a positive means of regulating 

the system flow and pressure. Pressure control valves were installed within the piping 

system to regulate the delivery pressure of the cleaning water to the required level. A 

pressure sustaining valve was installed on the main piping header to relieve excess flow 

and to maintain system pressure during times the full 3 m3

/min (800 gpm) flowed but 

was not required. 

Prototype testing occurred in conjunction with NYCDEP's normal basin cleaning 

operations. The basins were pumped and drained by plant staff prior to the normal 

cleaning operations. Plant staff cleaned the length of the basins, except for the 6.1 m 

(20 ft) test sections located in basins 1, 2, and 3 using the existing traveling bridge 

system. On occasion, some test sections were not available for testing either due to 

cleaning of the test section by plant staff or by inflow of the tide through leaking tide 

gates. Generally after the plant staff completed cleaning, the prototype system testing 

began. Sediment depth in the prototype system test sections was measured and 

recorded by entering the basins. The distribution of solids within each test section was 

documented. The tipping buckets, spray nozzles and orifice headers at the influent end 

of the basins were generally tested first. After the testing at the influent end of the 

basins was finished, the tipping buckets, flushing gate, spray nozzles and orifice 

headers at the effluent end of the basins were then tested. During several test runs, the 

prototype testing was halted early due to overloading of solids in the AWPCP’s cross 

collector channel and the chamber screen channels. As discussed later, sudden slugs 

of cleaning water, carrying large quantities of solids, overloaded the channels causing 

the testing to be ended prematurely. 

71
 
Figure 5.2.c-1. Wash Water Piping System Schematic 

72
 
As the results of the prototype testing were analyzed, the operating conditions were 

adjusted and the actual prototype system was modified or the entire prototype system 

was removed from testing when deemed ineffective. 

Tipping Buckets. Grande, Novac & Associates, Inc., of Montreal, Canada supplied the 

units tipping buckets. Four tipping buckets were installed in basin 1. Two buckets, one 

each on opposite sidewalls, were installed at elevation +1.5 m (+ 5.00 ft) at both the 

influent (TB-1A and TB-2A) and overflow (TB-1B and TB-2B) test sections of the basin, 

as shown in Figures 5.2.c-2 and 5.2.c-3. The buckets were installed at either end of the 

basin to test the cleaning effectiveness of the highest and lowest solids loading in the 

basin. The overflow end buckets were used to determine the ability of the buckets to 

convey the solids down the full length of the center velocity channel. 

Each bucket was 6.1 m (20 ft) long and had a liquid capacity of 0.15 m3 per linear meter 

(11 gal per linear foot [gal/ft]) for a total volume of 0.8 m3 (220 gal) per bucket. During 

testing, a constant supply of cleaning water was fed into the buckets. When the 

maximum water level was reached in the buckets, the buckets would automatically tip 

due to the change in the bucket's center of gravity, a 0.8 m3

 (220 gal) wave of water 

down the basin wall and along the floor, to the velocity channel. 

A training wall, stretching the width of the basin, from the velocity channel to the basin 

sidewall, was originally installed along the edge of the test area. The purpose of the 

training wall was to prevent the wave of cleaning water from dispersing laterally, thereby 

maintaining the full force of the wave within the test section. In addition, curved wall 

fillets were installed at the bottom of the sidewall to allow for a smooth transition of 

water flow from the wall to the floor. Figure 5.2.c-3 shows the approximate size and 

location of the training walls and wall fillets. 

Once the NYCDEP's cleaning operations were completed and the sediment in the 

basins measured, filling of the tipping buckets commenced. The two buckets at the 

influent end of the basins were filled simultaneously and then discharged automatically. 

The process was repeated as necessary until the test section was considered clean. 

Table 5.2.c-1 summarizes the test conditions for the tipping buckets along with the other 

prototype systems tested. 

After the installation of the equipment it was found that bucket TB-2B did not tip 

automatically. A counter weight was added to correct bucket mounting/misalignment 

and to permit the bucket to tip. Periodically, bucket TB-1B also experienced hesitation 

in tipping. 

Following the initial test runs, modifications were made to the training walls and fillets to 

improve their effectiveness at eliminating dead zones within the basins. The original 

training walls, made of steel angle approximately 15—30 cm (0.5—1 ft) in height, were 

bolted to the basin floor. The training walls extended from the walls at the column line 

to the velocity channel. During testing it was observed that the solids along the training 

73
 
Figure 5.2.c-2. Tipping Buckets – Enlarged Plan 

74
 
Figure 5.2.c-3. Tipping Buckets – Section A 

75
 
Table 5.2.c-1. Work Plan Prototype Test Matrix 

76
 
wall were the most difficult to clean. It was believed that this was due to a dead zone 

between the edge of the bucket and the basin wall pilaster. A modified plywood training 

wall was installed along the width of the pilaster to eliminate this dead zone as shown in 

Figure 5.2.c-4. 

In separate subsequent tests, the training wall from bucket TB-2A and the wall fillets 

from buckets TB-1B and TB-2B were removed. The training wall was removed to 

determine what effect its removal would have on the solids located in the dead zones. 

The wall fillets were removed for use in the modified orifice header configuration. 

Flushing Gate. The flushing gate system was tested for possible installation in the 

facility's influent barrels, and/or cross collector channels. Currently, accumulated solids 

in the influent barrels must be removed manually. The wave produced by a flushing 

gate can produce a high velocity flow of water over a relatively long distance. The basin 

velocity channel offered an opportunity to test the effectiveness of the flushing gate 

without actually entering the influent barrels. Grande, Novac & Associates, Inc. of 

Montreal, Canada supplied the flushing gate. A single flushing gate, 1.22 m (4 ft) wide 

with two 2.74 m (9 ft) long steel subwalls was installed at the overflow end of basin 1. 

The flushing gate was centered at the head of the existing velocity channel, while the 

two subwalls were located on either side of the flushing gate. The purpose of the 

subwalls was to accommodate the installation of the gate and to provide a storage 

volume of approximately 15.1 m3

 (4,000 gal) of flushing water. This system utilized both 

CSO and bay water as its source of flushing water. Cross sectional views of the 

flushing gate are shown in Figure 5.2.c-5. 

When the desired volume of water behind the flushing gate was obtained, a remote 

hydraulic actuator opened the flushing gate and sent a wave of cleaning water down the 

length of the velocity channel. The time of travel for the leading edge of the wave 

produced by the gate was measured at two points in the center velocity channel, 30.5 m 

(100 ft) and 76.2 m (250 ft) downstream of the gate. The velocity versus time of the 

flow following the wave was determined by measuring the time of travel of objects 

floating in the channel. 

There was no active means of closure for the flushing gate. After the stored wash-water 

volume had discharged, the flap portion of the gate would close under its own weight. 

The flap was required to sit flush against the gate frame in order for the hydraulic unit to 

lock the flap in the closed position. As a result, any debris or leaking tide water flow 

would interfere with the gate closure. When the gate closed, debris caught between the 

flap and frame would often cause leakage of the stored volume. Several times during 

the testing, a person was required to enter the basin in order to clear debris from the 

gate and allow it to be closed. On one occasion, an automobile tire became wedged in 

the gate opening. 

77
 
Figure 5.2.c-4. Modified Plywood Training Wall Schematic 

78
 
Figure 5.2.c-5. Flushing Gate – Section B 

79
 
Fixed Header with Spray Nozzles. Two separate spray header arrangements 

(Alternative-1 and Alternative-2) were installed in basin 2. Spraying Systems Inc. of 

Wheaton, Illinois provided spray nozzles installed on these headers. Alternative-1 used 

the model HH30100 nozzle that delivered a full cone spray pattern of 0.05 

m3

/min/nozzle (12.3 gpm/nozzle) at approximately 42,190 kg/m2 (60 psi). Alternative-2 

used the model PSS35200 nozzle that delivered a flat spray pattern of 0.1 

m3

/min/nozzle (24 gpm/nozzle) at approximately 42,190 kg/m2 (60 psi). Both 

arrangements also utilized a flat-spray-pattern-nozzle-model 50200 [delivering 

approximately 0.1 m3

/min/nozzle (24 gpm/nozzle) at 42,190 kg/m2 (60 psi) to clean the 

sidewalls. Each spray manifold was installed with 3 separate 10.2 cm (4 in.) header 

pipes. 

Two pipes were dedicated to clean the floor, and one pipe was dedicated for wall 

cleaning as shown in Figure 5.2.c-6. The spray water manifolds were located on 

opposite walls at both the influent and overflow ends of the basins at elevation +1.5 m 

(+5.00 ft). At this elevation the nozzles were approximately 4.3 m (14 ft) from the basin 

floor. This maintained the nozzles above the basin flood line and reduced the potential 

for nozzle damage and clogging. 

The flow rate and pressure of the cleaning water supplied to the nozzles were held 

constant throughout each test run. Provisions to adjust the flow and pressure, using 

throttling valves located in each header, were made so that these parameters could be 

varied from run to run. The cleaning system was allowed to run for up to 15 min. The 

flow rate and operating pressure were recorded for both nozzle arrangements. Table 

5.2.c-2 provides the test conditions for the fixed spray headers with nozzles. 

Even at the highest attainable pressures and flows, the nozzles proved ineffective at 

cleaning the basin floor and were removed from service. The nozzles were also subject 

to clogging from debris found in the cleaning water supplied from the spray water 

channel. 

Fixed Spray Header with Orifices. Initially, four headers were installed in two 

arrangements in the influent and effluent ends of basin 3. Two orifice headers were 

located on opposite walls at the influent end of the basin, and two orifice headers were 

located on opposite walls at the overflow end of the basin. Each orifice header was 6.1 

m (20 ft) long and equipped with 1.6 cm (5/8 in.) orifices spaced on approximately 30.5 

cm (1 ft) centers. The spacing and size of the holes were selected to provide a water 

flow rate of 0.5 m3/min/m (40 gpm/ft) at a discharge pressure of 7,030 kg/m2 (10 psi). 

One pair of headers was installed at a centerline elevation of +1.5 m (+5.00 ft) (influent 

header 1A and overflow header 2B) with the orifices aimed directly at the sidewall. 

These headers were designed to send cleaning water down the wall in a continuous 

sheet, thereby washing the basin floor and carrying solids to the velocity channel. The 

second pair of headers (influent header 2A and overflow header 2B) was installed at a 

centerline of elevation -8.50 with the header aimed directly at bottom of the wall. 

80
 
Figure 5.2.c-6. Spray Nozzle Header – Section 

81
 
Table 5.2.c-2. Summary of Prototype Testing Operation 

82
 
Table 5.2.c-2 page 2 

83
 
This system was also designed to provide a sheet flow of cleaning water along the floor. 

Since no nozzles were utilized under this alternative, there was less concern for 

clogging while the headers were submerged. Cross sectional views of this original 

orifice header configuration are shown in Figure 5.2.c-7. 

Training walls stretching the width of the basin from the velocity channel to the basin 

sidewalls were installed at the outside end of the basin test section. Similar to the 

tipping buckets, the purpose of the training walls was to prevent dispersion of the 

cleaning water and maintain the flow within the basin test section. 

After the initial six test runs, the original orifice header configuration was determined to 

be ineffective at cleaning. Subsequently, modifications to the header configuration were 

made including the following: 

•	 Removal of the overflow end orifice headers (for reuse in the modified configuration 

at the basin influent) 

•	 Rotation of the influent orifice header 1A (elevation +1.5 m [+ 5.00 ft]) to orient the 

water jets nearly tangential to the sidewall. 

•	 Rotation of the influent orifice header 2A to orient the spray downward and towards 

the velocity channel at a spray angle of approximately 30�s with the floor. 

•	 Installation of two new influent orifice headers (2B and 3) on the eastern side of the 

test section. Header 2B was installed on the basin floor at elevation –2.6 m [-8.5 ft] 

and header 3 was installed on the sidewall at elevation +1.5 m (+5.00 ft). The spray 

on header 2B was oriented similar to header 2A while header 3 was oriented similar 

to header 1A. 

• Installation of wall fillets at the wall-floor joints beneath influent headers 1A and 3. 

Figure 5.2.c-8 provides a section showing the modified header configuration. 

Flow rate, operating time and pressure was recorded for each orifice header. Table 

5.2.c-2 provides the operating conditions and performance for this system. 

Few operational and mechanical problems were encountered while operating this 

system. Occasionally rags and other debris became entangled around the supports of 

the orifice headers; however, this debris did not effect the cleaning operation. 

Submersible Mixer. A single submersible 18.6 kW (25 HP) jet mixer was supplied by 

Davis EW of Thomasville, Georgia (now a division of US Filter). The jet mixer was 

installed at the influent end of Basin No. 2 using steel guide rails and lifting cables 

similar to those typically associated with submersible pumping equipment. The mixer 

was provided with an integral diffuser and air supply piping system to achieve a roll 

mixing effect within the basin. 

84
 
Figure 5.2.c-7. Original Orifice Header Configuration – Section A 

85
 
Figure 5.2.c-8. Modified Orifice Header Configuration – Section A 

86
 
The jet-mixer was placed into continuous service after test run no. 3. The mixer was 

controlled using automatic level control floats to turn on while the basins were filling 

during a CSO event and off when the basins were almost completely drained. The 

mixer remained in operation sometimes for more than one day until the basin was 

drained. 

The purpose of the mixer was to provide re-suspension of the solids, such that solids 

would be removed with the wastewater during the de-watering operation, minimizing 

solids deposition within the basins. During testing the mixer proved ineffective at resuspending the solids in the test area and reducing solids deposition. The mixer was 

removed from service after test run no. 7. Following removal of the jet mixer from 

service, it was found that the diffuser and discharge piping of the mixer were almost 

completely clogged with leaves, rags and other debris. 

Prototype Test Results 

Solids Composition and Distribution. As would be expected for a CSO facility, the 

composition and distribution of residual solids in the detention basins varied significantly 

between events. The measured depth of solids ranged from a trace coating (less than 

1.5 cm) up to 45.7 cm (1.50 ft) at individual piles. The solids ranged in consistency from 

a thin watery mixture, to a viscous “clay like” consistency, to heavy grit. Test events 

with the greatest depth of solids and heavier grit generally followed rainstorms of high 

intensity. Test events with fewer solids and solids with a thinner consistency generally 

followed rainstorms with low intensity or low total precipitation. 

In general, the influent ends of the basins contained the heavier solids, with the overflow 

ends characterized primarily by floatables, leaves, and a thin coating of solids. The 

solids depth was greater at the influent end than at the overflow end. Dead zones at the 

influent headwall of each basin contributed to the formation of piles or long mounds of 

solids in the influent test sections. The largest piles were observed on the basin side of 

the influent headwall columns and along the sidewall. Solids appeared to build up 

behind these obstructions, as scouring of solids during CSO inflow was prevented. 

These and other piles of solids often posed the most resistance to cleaning. Within 

basins no. 1 and 3, a ridge of deeper solids was observed several times along the 

western wall of the basin influent. At the influent end of basin No. 1, the solids depth 

was generally greater on the western side of the test section than the eastern side. This 

likely contributed to the formation of piles along the western sidewall. The difference in 

solids depth between the two sides of the basin No. 1 influent may have been due to a 

partial blockage of the influent barrel feeding the eastern side of this basin, resulting in 

unequal flow distribution. 

Based on subjective observations, it appeared that the consistency of the residual solids 

was generally thicker and more cohesive when more days had transpired between the 

inflow event and the cleaning test as compared to when cleaning was performed 

immediately following an inflow. The nature of the solids consistency affected cleaning 

system performance, with the thicker/ cohesive solids being more resistant to cleaning. 

87
 
Benchmark of Existing Cleaning Operations. As discussed previously, the existing 

traveling bridge cleaning system operations were observed to document general 

operations, cleaning time, and water consumption. The cleaning time, water 

consumption and estimated electrical cost for this system are presented in Table 5.2.c3. 

The bridges provide very effective cleaning of the basin walls and floor. Generally, 

several complete passes along the basin length were required to clean each basin. 

With the bridge mounted water cannon, solids or objects which might have been 

unaffected by the spray nozzles were readily moved into the velocity channel and 

cleaned from the basin floor. The bridges readily achieved complete basin cleaning. 

The primary advantage of the traveling bridge system is that it is very effective at 

cleaning the basins. The degree of cleaning is only limited by the amount of time an 

operator expends in cleaning. Disadvantages observed with the bridge system included 

the following: 

•	 Cleaning with the bridges was labor intensive, requiring staffing of each bridge 

during operation 

•	 Because there is no separation from the basin airspace, the operator had the 

potential to be exposed to H2S levels exceeding standards 

•	 The bridges appear to require frequent maintenance. Often during the test runs, one 

of the six bridges was observed to be temporarily out of service, due to problems 

with the bridge controls or with the electrical bus bar supplying each bridge 

Tipping Buckets. The test results for the tipping bucket prototype cleaning system are 

presented in Table 5.2.c-4. This table presents tabulated results for what are 

considered representative solids conditions. For several test events, the preceding 

inflow did not result in any significant deposition of solids, or the test section may have 

been cleaned by tidal inflow or by plant personnel. The data for such events was not 

considered representative and is not included in Table 5.2.c-4. Data for the influent and 

overflow test sections are presented separately, with averages for the number of bucket 

tips per test section and projected water usage per basin calculated over each test run, 

each test section (influent, overflow), and over the entire basin. The average over the 

basin length assumes a linear distribution of solids (and associated cleaning water 

consumption) between the solids depth measured at the influent test section, and that 

measured at the overflow test section. 

The tipping buckets provided very effective cleaning of the sidewalls and the basin floor. 

The high velocity wave produced by each bucket tip easily cleaned solids and debris 

from the basin floor and conveyed them to the center velocity channel. With greater 

depths of solids or for heavier solids or grit, more tips were required as compared to test 

conditions with fewer solids or with looser solids. 

88
 
Table 5.2.c-3. Operating Data for Existing Traveling Bridge 

89
 
Table 5.2.c-4. Tipping Bucket Cleaning System Test Results at Representative Solids 

Conditions 

90
 
Averaged over the basin length, 7 tips are estimated to be required per bucket, resulting 

in a projected water usage of 291.4 m3

 (77,000 gal) per basin. The range in projected 

water usage is from a low of 41.6 m3 (11,000 gal) per basin, to greater than 1249 m3 

(330,000 gal) per basin. The highest projected water consumption was required to 

clean a pile of matted leaves mixed with solids in the influent test section during test run 

no. 16. In this case, the leaves seemed to bind the solids together, resulting in a single 

large mass that was difficult to clean. This pile was slowly eroded with each tip. After 

30 tips the test was terminated while the pile still remained. Difficulty cleaning leaves 

was also observed in the overflow test section of run no. 10. There was a pile of matted 

leaves in the test section for bucket TB-1B. With each tip, the entire mass of leaves 

was lifted, carried a short distance, and redeposited. Ultimately, 12 tips were required 

to wash this pile into the center velocity channel. 

The solids that had collected along the training wall proved difficult to clean and were 

often the limiting factor in determining when the basin was 99% clean. Elsewhere on 

the test section floor, the wave of wash-water was able to mix and resuspend the solids 

in the wash-water. However, along the training wall the solids would generally erode 

slowly with each tip. In the prototype installation, the wall-floor fillet on the eastern tank 

wall was installed flush against the sidewall in the recess between the column pilasters. 

As a result, over the pilaster width, there was no smooth transition for the wave between 

the wall and floor, and the wave would lose energy in this dead zone when it impacted 

the floor. The modified plywood training-wall improved the situation somewhat by 

eliminating the dead zone created by the sidewall pilaster. The plywood training-wall 

worked by preventing the deposition of solids in the area blocked by the pilaster. Solids 

along the modified training wall still eroded slowly with each tip. However, the amount 

of solids collected along the training wall was less than for the original vertical trainingwall installation. 

When the training-wall was removed from the test area for bucket TB-2A, the cleaning 

effectiveness was the same if not more effective than with the training-wall. Without the 

training-wall the wave could penetrate laterally into the solids, resulting in more effective 

resuspension of the solids at this location. 

Operationally, the buckets were very simple to operate, requiring only that the water be 

turned on; however, difficulty in tipping first one bucket (TB-2B) then another bucket 

(TB-1B) was observed. Due to possible misalignment of TB-2B during installation, 

counterweights were added to the bucket so that it would tip automatically during filling. 

This worked for a time. However, later in the test program the bucket had to again be 

tipped manually. Periodically, the same problem was observed with bucket TB-1B. 

Flushing Gate. The flushing gate produced a high velocity flow in the center velocity 

channel that extended at least 76.2 m (250 ft) down the length of the tank. The velocity 

of the leading edge of the wave is a function of the slope of the channel and the steep 

hydraulic grade created by the wave surge. The center velocity channel slopes 

approximately 0.63 percent towards the cross channel. The flow velocity was 

91
 
calculated using Mannings Equation ranged between 1.3—1.9 m/s (4.4—6.2 ft/s) for 

uniform flow at typical wash-water depths of 0.15 and 0.30 m (6 and 12 in). During 

operation of the flushing gate, the leading wave velocity was measured at 3.4—4.3 m/s. 

(11-14 ft/s). The water velocity in the channel later dissipated to 1.2 m/s (4 ft/s) after 

approximately 40 s and 100 s at the 30.5 m and 76.2 m (100 ft and 250 ft) measuring 

stations, respectively. Although there were generally no significant solids within the 

velocity channel, this high initial flow rate would be expected to scour most if not all 

solids and objects from the channel. 

Based on the size and configuration of the gate supplied, which was primarily designed 

to flush the velocity channel, the lateral extent of the high velocity flow was limited to the 

velocity channel. In the immediate vicinity of the gate, the high velocity flow extended 

laterally approximately 61—152.4 cm (2—5 ft) from the center channel. However, the 

high rate of discharge produced backwater within the basin effluent test area, which did 

not result in any effective cleaning of the test area outside of the velocity channel. 

As indicated previously, significant operating problems were encountered with the 

flushing gate. These problems were primarily related to floatables and debris that 

caused problems with the closure and sealing of the gate. 

Fixed Header with Spray Nozzles. The test data for the spray nozzle prototype system 

are presented in Table 5.2.c-5. The spray nozzles did not provide effective cleaning. 

The nozzles only cleaned the immediate area impacted by the central portion of the 

spray jet. Outside the central portion or cone of the spray jet, the nozzles produced a 

fine mist, which was ineffective at cleaning. Away from this immediate area the washwater would form channels in the deposited solids and would not provide any further 

cleaning. Overall, approximately only 25% of the floor test-area was cleaned by either 

nozzle configuration. This degree of cleaning was achieved within the first minute of 

operation. The system was generally allowed to operate for 15 min. However, the 

added operating time did not result in any further cleaning. Since the spray nozzles did 

not provide complete cleaning, the cleaning times and volumes presented in Table 

5.2.c-5 have been qualified to indicate that greater cleaning time and volumes would be 

required. The flat spray nozzles directed at the sidewall provided an adequate job of 

cleaning the wall only. 

Based upon field observations, it appears that in order to provide adequate cleaning 

with this spray nozzle configuration, the headers would need to be swept back and forth 

across the floor during operation, similar to the existing traveling bridge operation. 

Fixed Header with Orifices. The test data for the original and modified orifice header 

configurations are presented in Tables 5.2.c-6 and 5.2.c-7, respectively. The original 

header configurations provided only partial cleaning of the floor test area. An area 

extending approximately 1.5—3 m (5—10 ft) from the sidewall was completely cleaned 

within the first minute of operation. These solids were usually cleaned with the initial 

surge of wash-water. 

92
 
Table 5.2.c-5. Spray Nozzle Cleaning System Test Results (2 pg.) 

93
 
Table 5.2.c-5. Spray Nozzle Cleaning System Test Results (2nd pg) 

94
 
Table 5.2.c-6. Original Orifice Header Configuration Test Results 

95
 
Table 5.2.c-7. Modified Orifice Header Configuration Test Results at Representative 

Solids Concentrations (3 pg.) 

96
 
Table 5.2.c-7. Modified Orifice Header Configuration Test Results at Representative 

Solids Concentrations 2nd page 

97
 
Table 5.2.c-7. Modified Orifice Header Configuration Test Results at Representative 

Solids Concentrations (3rd page) 

98
 
Further away from the wall, the wash-water would form channels and the remaining 

floor area would be cleaned only very slowly, as the flow eroded the edges of the 

channel. Based upon field observations, it appeared that a higher velocity of washwater was required to completely resuspend the solids over the full width of the test 

section and resist the formation of channels. The headers installed at elevation +1.5 m 

(+5.00 ft) provided adequate cleaning of the sidewall. The system was generally 

allowed to operate for 15 min, by which time only 30—75% of the floor test area was 

effectively cleaned. Since the cleaning was incomplete, the cleaning times and volumes 

shown in the table are qualified to indicate that actual values are greater than those 

shown. 

Overall, the modified orifice header configuration showed significant improvement over 

the original orifice header configuration. The modified orifice header no. 1A provided 

the same degree of cleaning, approximately 30—75% of the floor test area, as the 

original configuration. However, the area appeared to be cleaned sooner than the 

original configuration. Additionally, the wall-floor fillet minimized the quantity of solids 

deposited along the sidewall. While no significant solids/grease buildup was observed 

on the basin sidewalls, the spray from modified header no. 1A provided better cleaning 

action than original configuration. Because of the tangential flow spray of the washwater onto the wall, all of the water flow washed down the wall in a sheet of water. In 

the original configuration in which the orifices were directed more at the wall, a large 

portion of the flow sprayed off the wall and rained down on the basin floor. 

The combination of modified headers no. 2A, 2B, and 3 provided complete and effective 

cleaning of the test area floor and sidewall within a short period of time. The 

combination of the two lower headers (2A and 2B) spraying towards the velocity 

channel provided a high scouring velocity and prevented the formation of channels in 

the flow path that would otherwise limit cleaning effectiveness. With all three headers 

operating, the test area was generally 95% clean within the first minute of operation. As 

shown in Table 5.2.c-7, the average required cleaning time for the test section was 2.68 

min with a projected basin wash-water usage of 530 m3/basin (140,000 gal/basin). 

During the prototype testing, the header operating sequence and flow rates were varied 

to determine the optimum operating conditions. Simultaneous operation of headers 2A 

and 3, followed by header 2B alone, appeared to minimize the consumption of washwater. Simultaneous operation of all three headers was also tested, as this was most 

representative of the future full-scale operating scenario. This provided cleaning 

effectiveness equal to sequential operation in a shorter time period, but at higher 

projected water consumption. A wash-water flow rate of 2.3 m3/min (600 gpm) per 

header (30 gpm/If /header) was sufficient to clean the majority of the solids. However, 

in test run no. 13, a partially full trash bag was present in the test area and a flow rate of 

2.8 m3

/min (750 gpm) per header (37.5 gpm/If /header) was required to move the trash 

bag into the velocity channel. The purpose of header no. 3 was to clean the sidewall 

and the area behind header no. 2A. The flow rate on header number 3 was varied 

between 0.8—1.5 m3/min (200—400 gpm or 10—20 gpm/ft). A flow rate of between 

99
 
0.8—1.2 m3/min (200—300 gpm) appeared to be sufficient to meet these cleaning
 

objectives.
 

Operation of the orifice headers was simple. The biggest difficulty was modulating the
 

flow at startup to minimize the occurrence of an initial surge of water. The steps taken
 

to minimize this effect included slowly opening the valves at startup and tapping the
 

feed pipe to provide air relief points. During operation, some rags and debris became
 

entangled on the orifice header pipe supports. The amount of material caught on the
 

supports was minor and did not affect cleaning performance. It is important to note that
 

after a total of 6 months of testing the lower headers, no clogging of orifices was
 

observed. This was originally identified as a concern for submerged orifices.
 

Submersible Mixer. There is no specific test data on the submersible mixer other than
 

solids depth measurements and qualitative observations. From observations of the
 

residual solids distribution following basin de-watering, the mixer did not appear to
 

effectively minimize the deposition of solids. Typically, solids were scoured from a small
 

area approximately 1.5—3.0 m (5—10 ft) in diameter immediately at the discharge of
 

each of the mixer nozzles. However, the solids depth immediately surrounding the
 

scoured areas was deeper than for other basins, indicating that the scoured solids were
 

immediately redeposited away from the nozzle discharge. Often there was a deep pile
 

or ridge of solids formed around the scoured area. Elsewhere on the test area floor, the
 

solids were unaffected by the mixer.
 

The mixer did not have any effect on floatables and leaves, which remained in the test
 

area following basin de-watering.
 

The discharge nozzles would periodically become clogged as indicated by decreased
 

mixing turbulence in the tank during mixer operation. Following removal of the
 

submersible mixer prototype, the flow distribution box and several of the nozzles and
 

feed lines were found to be clogged with leaves and debris. This material apparently
 

became so compacted in the distribution box that the top weld of the box cracked open.
 

Comparison of Technologies and Recommendations for Design
 

Summary of Cleaning Effectiveness. The results of CDM’s field testing program
 

demonstrated that either the tipping buckets or the modified orifice header system
 

would provide satisfactory cleaning of the basin floor, equivalent in performance to the
 

existing traveling bridge system. Each of these two systems achieved complete
 

cleaning of the test areas.
 

The spray headers equipped with nozzles and the original orifice header configuration
 

provided only partial cleaning of the test areas. This configuration was considered to be
 

unsuitable for application at Spring Creek given the existing basin geometry. The spray
 

headers only provided cleaning of the immediate area impacted by the nozzle spray.
 

This resulted in cleaning only 25% of the test area. The original orifice header
 

configuration was designed to provide sheet flow of water over the basin floor. With this
 

100
 
system, channeling of the flow occurred rapidly, which limited the ability of the system to 

clean the entire test area. Under the best of conditions, this system cleaned only up to 

75% of the test area. These two systems are, therefore, not considered further. The 

submersible mixer did not minimize the deposition of solids in the test area and created 

deep piles and ridges of solids adjacent to scoured areas. Also, the mixer became 

clogged with leaves and debris. Therefore, the submersible mixer is not considered 

further. 

The flushing gate was effective at producing a high velocity wave with initial velocity of 

3.4—4.3 m (11—14 ft) per second in the center velocity channel at 30.5—76.2 m (100— 

250 ft) downstream of the gate. This velocity would be more than sufficient to scour 

solids from the velocity channel or the influent barrels. It was not intended to evaluate 

the flushing gate for cleaning of the basin floor. The gate was demonstrated on the 

center velocity channel as a surrogate for possible application to the influent barrels. 

Operational problems were encountered with floatables and debris interfering with gate 

closure and sealing. For full-scale application, some means of positive closure would 

be recommended. Also, if an alternate method of sealing without significant leakage 

could not be developed, an alternate water supply would be required to fill the storage 

volume prior to a flushing event. 

The remainder of the discussion in this section will address the two successful prototype 

systems for basin cleaning, tipping buckets and orifice headers. 

The average cleaning requirements determined from the prototype testing for the tipping 

bucket and modified orifice header systems are summarized in Table 5.2.c-8 and are 

compared against the existing traveling bridge system. The cleaning requirements for 

the tipping buckets and orifice headers are based upon cleaning a single 6.1 m (2 ft) 

length of the basin at one time, which was the length of the test areas. The cleaning 

time indicated for both the buckets and orifice headers could be reduced from that 

shown in the table by increasing the size of the cleaning zone, and consequently, 

increasing wash-water pumping and conveying requirements. However, for the purpose 

of comparing the prototype test data, the cleaning requirements for the two systems are 

presented on the same 6 .1 m (20 ft) basis. 

As shown in the Table 5.2.c-8, the average basin cleaning time for the tipping buckets is 

more than twice that for either the orifice headers or traveling bridges. This is primarily 

due to the multiple tips required to clean each zone. For full-scale implementation of 

either the tipping buckets or orifice headers, the actual time to clean all six basins would 

not be six times the average basin cleaning time, but rather some fraction of this 

quantity. The total cleaning time would be dependent upon a number of factors 

including: 

• Length and number of cleaning zones in each basin 

• Number of basins to be cleaned simultaneously 

• The maximum allowable or cost-effective wash-water flow rate 

101
 
• The maximum allowable solids loading rate to the cross channel and climber 

screens 

The last two factors would limit the number of cleaning zones in operation at any one 

time. The flow and solids loading rates selected would impact the sizing and cost of the 

cleaning and de-watering screening equipment. Therefore, while it is difficult to assess 

the minimum full-scale cleaning time without considering all the above factors, it is 

believed that the use of the tipping buckets will result in a significantly greater total 

cleaning time than either orifice headers or traveling bridges. 

Table 5.2.c-8. Projected Cleaning Requirements for a Single Basin Cleaning Event 

(Based on 20-ft. Cleaning Zone). 

Mod. Orifice Traveling Cleaning Parameter Tipping Headers (2A, 2B, Buckets Bridge and 3) 

Average Time (min/basin) 175 67 72 

No. of Cycles/Zone (averaged 7 tips 1 2-6 passes 

over basin length) 

Water Consumption 

- Influent Test Area 	 (m3/basin) 457.99 530.77 NA 

(gal/basin) 121,000 140,229 NA 

- Average Entire Basin (m3/basin)	 291.45 NA 272.05 

(gal/basin) 77,000 NA 71,875 

Estimated Electrical Consumption 

(kW-hr/basin) 40 40 40 

Level of Staffing Required for Automated Automated Fully Manned 

Operation 

Note: NA denotes data not available. 

The performance of the tipping buckets and the modified orifice headers compare 

favorably with each other and with the traveling bridge system in that all three systems 

generally are capable of achieving complete cleaning of the basin floor and walls. In 

one instance, run no. 16, the tipping buckets could not completely clean a pile of matted 

leaves mixed with solids. After 30 tips the test was terminated as it was determined that 

no further cleaning could be achieved. Similarly, difficulty with the tipping buckets 

moving leaves was also noted in the overflow test area of run no. 10, where 12 bucket 

tips were required to move a pile of leaves. The orifice headers exhibited similar 

difficulty moving a trash bag. However, once the lower header flow rate was increased 

to 2.8 m3

/min (750 gpm) per header, the bag was swept into the velocity channel. With 

102
 
the proper design flow rate, the orifice headers should be capable of providing complete 

cleaning. 

The average wash-water consumption for the tipping buckets and the traveling bridges 

is similar at 270—290 m3 (71,875—77,000 gal) per basin. There is no average over the 

entire basin length for the modified orifice headers as this system was only tested at the 

basin influent end. This represents the worst case solids cleaning condition. For the 

influent test areas, projected basin water consumption values are presented for the 

tipping buckets and modified orifice headers. Although, based on this test area the 

tipping buckets use approximately 14% less water than the orifice headers. The washwater volumes ranged from 100 m3

 to greater than 1,250 m3 (27,500 to >330,000 gal) 

per basin for the tipping buckets and 175—880 m3

 (46,050—232,900 gal) per basin for 

the modified orifice headers. This data variability would need to be accounted for in fullscale design by assuming conservative design values. Therefore, the difference in 

wash-water consumption between these two systems is not considered significant. 

Estimates of power consumption for the three systems are presented primarily for 

informational purposes. Because the cleaning systems are operated only intermittently, 

on the order of 8 h/week, the differences in electrical costs are insignificant. As shown 

in the table, the tipping buckets are estimated to use the least amount of electricity while 

the orifice headers use the most. This is because the orifice headers are a high flow 

rate, moderate head system, while the buckets are a low flow and low head system. 

Both the tipping buckets and orifice header system would be designed for automatic 

operation in full-scale application. This is an advantage over the traveling bridges that 

are continually manned during operation. 

Cost Comparison 

The estimated cost of implementing either tipping buckets or orifice headers at the 

Spring Creek facility are presented in Table 5.2.c-9. These estimates are based upon 

average solids conditions and water consumption. For the tipping buckets, water 

consumption was averaged over the length of each basin based upon the influent and 

overflow test area results. The modified orifice header system was only tested on the 

basin influent end. Similar to the tipping buckets, the water consumption for the 

overflow end of the basin would be expected to be less than the influent end, due to the 

distribution of solids. For this analysis, this was not taken into account. It is likely that 

the operating time for the orifice headers, and associated electrical cost, could be than 

the calculated electrical cost. 

103
 
Table 5.2.c-9. Cost Comparison of Full Scale Tipping Bucket Vs. Orifice Header 

Cleaning Systems 

Item Tipping 

Buckets 

Orifice Headers 

Piping Alt 1 

Pipe Gallery 

Piping Alt 2 

Pipe on Roof 

Deck 

Direct Capital Costs 

Piping and Valves $ 781,000 $ 10,255,107 $ 5,840,000

Mechanical Equipment $ 4,956,000 $ 120,000 $ 120,000

Instrumentation and Controls $ 10,000 $ 16,000 $ 16,000

Electrical Equipment $ 362,000 $ 300,000 $ 300,000

Indirect Capital Costs 

Contractor OH&P (1) $ 1,283,000 $ 2,245,000 $ 1,318,000

Design / Contingency (2) $ 916,000 $ 1,069,000 $ 941,000

SUBTOTAL CAPITAL COSTS $ 8,308,000 $ 14,005,107 $ 8,535,000

Annual O&M Costs 

Electrical Costs $ 1,500 $ 3,000 $ 3,000

SUBTOTAL O&M COSTS $ 1,500 $ 3,000 $ 3,000

O&M PRESENT WORTH (3) $ 17,205 $ 34,410 $ 34,410

TOTAL $ 8,325,205 $ 14,039,517 $ 8,569,410

Notes: 

1. Contractor's overhead and profit are assumed to be 21% of the direct capital cost. 

2. Contingency is assumed to be 15% of the direct capital cost for the Tipping Buckets 

and Orifice Headers Alt 2, and 10% for Orifice Headers Alt 1. 

3. Present worth analysis assumes a 20-year design life and an interest rate of 6.0%. 

The design assumptions used in this cost analysis were as follows: 

Tipping Bucket Design Assumptions 

Bucket unit volume
 2.7 m3

/m (220 gal/ft)
 

Bucket length
 6.1 m (20 ft)
 

Number of buckets/basin
 

Average water consumption
 

50
 

290 m3 (77,000 gal)/basin
 

Material of construction
 316L Stainless steel
 

Orifice Header Design Assumptions 

Number of headers/zone 3
 

Length of cleaning zone 24.4 m (80 ft)
 

Orifice header diameter 

Orifice pressure drop 

25.4 cm (10 in.)
 

7030 kg/m2

 (10 psi)
 

104
 
Cleaning flow rate (for each side of basin)
 

Wall header 0.2 m3

/min/m (15 gpm/lin ft)
 

Floor headers 0.5 m3/min/m (40 gpm/lin ft) /header
 

Total cleaning flow rate 1.2 m3

/min/m (95 gpm/lin ft)
 

Average water consumption 530 m3 (140,229 gal) /basin
 

Conceptual level capital and electrical operating costs are presented. The total net 

present worth cost is calculated based upon design life of 20 years and an interest rate 

of 6.0%. Costs are presented for two different feed pipe layouts for the orifice header 

system. Alternative-I is based upon installation of the main wash-water feed headers in 

pipe galleries located in the former spray water channels. This alternative has been 

included in the 30% design phase. Due to the space constraints in the gallery and 

obstructions by support columns, this alternative involved significantly more fittings and 

pipe flanges than the second alternative. Alternative-2, the preferred alternative for the 

orifice headers, is based on installation of the main wash-water feed headers on top of 

the basin cover system. Without the obstructions of the pipe gallery, the feed headers 

could be installed in straight pipe runs with pre-fabricated sections and with branches 

running 90� off the main header. This significantly reduced the number of fittings and 

pipe flanges. For comparison to the tipping buckets, only the cost for alternative-2 will 

be considered. 

As shown in Table 5.2.c-9, the estimated present worth costs to implement either the 

tipping bucket cleaning system or the orifice header cleaning system are equivalent. 

There is only approximately a 3% difference in these estimates, with the tipping bucket 

cost being slightly less. The operating electrical costs are not significant compared to 

the total capital cost. The cost for regular maintenance of the equipment is not included 

in the estimate. The primary maintenance items could include the following: 

• Maintenance and periodic replacement of pumps and valves 

• Maintenance and periodic replacement of tipping bucket bearings 

• Repair/replacement of orifice headers 

• Entry into basins to remove large debris items 

Currently long-term (e.g., 10 or 20 year) maintenance data is not available on tipping 

bucket or orifice header installations in CSO facilities. It would be difficult to assign 

maintenance/ replacement costs for servicing of the bucket bearings or the orifice 

header pipes. Also, the cost for staffing during cleaning operations has not been 

included in the cost comparison. Both the tipping bucket and orifice header systems 

would be automated for full-scale application. Some minimal staffing would be required 

to monitor and troubleshoot the cleaning operations. However, this could largely be 

combined with other facility staffing responsibilities. A staffing labor estimate for either 

of these two systems would be dependent upon operating philosophy and configuration 

(e.g., sequential vs. concurrent basin cleaning, staffing level). It is expected that the 

staffing cost for these two alternatives would be similar and would be significantly less 

than the costs for the existing traveling bridges. Assumptions on the extent of O&M 

costs required could significantly affect the total costs. Given this uncertainty on the 

105
 
O&M costs, only the electrical costs which are well defined have been included in the 

total present worth cost. 

Implementability and Operability 

Both cleaning technologies, tipping buckets and orifice headers, are readily 

implementable. There are no known constructability issues that would limit the 

applicability of one technology over the other. Both technologies provide effective 

cleaning. However, matted leaves posed a cleaning problem to the tipping buckets 

while this was not observed to be an issue with the orifice headers. Because of the 

higher velocity imparted by the orifice headers, this technology appears to be better 

suited at cleaning matted leaves, piles, and large objects. 

There were two operability issues noted with the tipping buckets, including potential 

bucket misalignment and transient overloading of the cross channel and climber 

screens by waves carrying high solids loads. Misalignment of one of the test buckets 

(TB-2B) caused it not to tip automatically. This was retrofitted in the field by adding 

counterweights. Later a second bucket (TB-1B) developed a hesitation in tipping 

automatically. With up to 300 buckets planned to be installed in a full-scale facility, this 

would present a significant concern. The second issue was a concern whenever the 

facility received a high inflow of grit. During testing, high quantities of grit were washed 

into the cross channel with the initial few tips of the buckets. Because the buckets 

produced discontinuous flow surges separated by zero flow conditions, the grit would 

often settle out in the channel and immediately before the climber screen when the 

water velocity decreased. This caused the channel and climber screen to become 

clogged with grit, which then had to be removed manually. 

With the orifice headers there are two operability concerns, entanglement of debris on 

the pipe supports and potential clogging of the orifices. Regarding the first issue, some 

rags/debris became entangled on the pipe supports of the prototype system during 

testing. However, the spray from the orifices was sufficient to prevent a buildup of 

solids with the rags. With proper design, the number of pipe supports and their 

configuration can be selected so as to minimize this effect. Periodic entry of the basins 

will likely be required to remove entangled material. As for potential clogging of the 

orifices, no clogging of the prototype system was observed during the 6 months the 

headers were in the basins. If clogging does occur, the orifices in the PVC headers 

may be easily cleaned by plant staff. 

106
 
5.3 Converting Dry Ponds for Enhanced Treatment 

Dry ponds are extensively used throughout the U.S. and other countries (EPA 1983). 

These ponds have been constructed to reduce peak runoff rates (peak shaving), 

typically with little consideration to runoff quality improvement. Dry ponds have been 

widely used in flood control projects to detain runoff and thus reduce peak flows and 

water elevations in the receiving waters. These flow reductions can also improve the 

aquatic habitat by reducing flushing of fish and other organisms from urban creeks (Pitt 

and Bissonnette 1984). Flow reductions also reduce downstream channel bank erosion 

and bottom scour. The use of many dry ponds in a watershed, without regard to their 

cumulative effect, can actually increase downstream flooding or channel scour problems 

(McCuen, et al. 1984). The delayed discharge of a mass of water from a dry pond may 

be superimposed on a more critical portion of the receiving water hydrograph. 

Because these ponds are normally dry and only contain water for relatively short 

periods of time, they can be constructed as part of parking lots, athletic fields, tennis 

courts and other multi-use areas. The outlets are designed to transmit all flows up to a 

specific design flow rate, after which excess flows are temporarily backed-up. In many 

cases, the ponds only contain water during a few rains each year. 

Several dry detention ponds were examined as part of the National Urban Runoff 

Program (NURP), with monitored pollutant removals ranging from insignificant to quite 

poor (EPA 1983). Sedimentation may occur in dry ponds, but only during the major 

storms when flows are retained in the pond. The deposited material must be removed 

after each treated rain, or it can easily be re-suspended by later rains and washed into 

the receiving waters. Adler (1981) found that new sediment deposits have little cohesion 

and without removal as part of a maintenance program, or without several feet of 

overlaying water, bottom scour is probable. Because of the poorly documented 

stormwater pollutant control effectiveness of dry detention ponds, they cannot, by 

themselves, be recommended as viable water quality control measures. However, they 

can be very effective when used in conjunction with other stormwater control practices 

(e.g., between a wet detention pond and an infiltration or grass filter area). 

Stanley (1996) examined the pollution removal performance at a dry detention pond in 

Greenville, NC, during eight storms. The pond was 0.7 ha in size and the watershed 

was 81 ha of mostly medium density single family residential homes, with some 

multifamily units, and a short commercial strip. The observed reductions were low to 

moderate for SS (42—83%), phosphate (-5—36%), nitrate nitrogen (-52—21%), 

ammonia nitrogen (-66—43%), copper (11—54%), lead (2—79%), and zinc (6—38%). 

Stanley (1996) also summarized the median concentration reductions at dry detention 

ponds studied by others. In all cases, the removal of the stormwater pollutants was 

substantially less than would occur at well designed and operated wet detention ponds. 

The re-suspension of previously deposited sediment during subsequent rains was 

typically noted as the likely cause of these low removals. The conditions at one of the 

ponds in Greenville, NC were observed three years after its construction. The most 

notable change was that the pond bottom and interior banks of the perimeter dike were 

107
 
covered with weeds and many sapling trees (mostly willows), which indicated that the 

interior areas were too wet to be mowed. The perforated riser was also partially 

clogged and some pooling was occurring near the pond outlet. It appeared that the dry 

pond was evolving into a wetland. The monitoring activity had been conducted a few 

months after the pond was constructed and was not affected by these changes. Stanley 

felt that the wetlands environment, with the woody vegetation, if allowed to spread, 

could actually increase the pollutant trapping performance of the facility. With continued 

lack of maintenance, the dry pond will eventually turn into a wet pond, with a significant 

permanent pool. The pollutant retention capability would increase, at the expense of 

decreased hydraulic benefits and less flood protection than originally planned. 

Nix and Durrans (1996) examined the benefits of off-line stormwater detention ponds. 

Off-line ponds (side-stream ponds) are designed so that only the peak portion of a 

stream flow is diverted to the pond (by an in-stream diversion structure). They are 

designed to reduce the peak flows from developed areas, with no direct water quality 

benefits, and are typically dry ponds. Off-line ponds are smaller (by as much as 20— 

50%) than on-line ponds (where the complete storm flow passes through the pond) for 

the same peak flow reductions. However, the outflow hydrographs from the two types of 

ponds are substantially different. The off-line ponds produce peak outflows earlier and 

the peak flows do not occur for as long a period of time as on-line ponds. If located in 

the upper portion of a watershed, off-line ponds may worsen flooding problems further 

downstream, whereas downstream on-line ponds tend to worsen basin outlet area 

flooding. Off-line dry ponds can be used in conjunction with on-line wet ponds to provide 

both water quality and flood prevention benefits. Off-line ponds have an advantage in 

that they do not interfere with the passage of fish and other wildlife and they do not 

dramatically affect the physical character of the by-passed stream itself. On-line dry 

ponds may substantially degrade the steam habitat by removing cover and radically 

changing the channel dimensions. The peak flow rate created by online dry ponds 

reductions can also have significant bank erosion benefits in the vicinity of the pond, 

although these benefits would be decreased further downstream. 

Case Studies 

The retrofitting detention pond case studies presented in this section cover a broad 

range of conditions, from underground facilities for very small paved drainage areas, to 

large detention facilities for major drainage areas. The problems encountered in each 

case history are described, along with the costs and benefits, to the extent that 

information was available. 

5.3.a. Sunnyvale Detention Basin, Santa Clara County, California 

South San Francisco Bay has a serious heavy metal pollution problem, especially for 

copper, and numerous methods are being investigated to reduce the discharges of 

metals to the Bay. Woodward Clyde Consultants (1994) conducted a retrofit project for 

the Santa Clara Valley Nonpoint Source Pollution Control Program to demonstrate the 

108
 
benefits of modifying an existing dry detention pond for enhanced water quality benefits. 

The discussion in this section is mostly taken from that report. 

According to a survey conducted in 1990 by Woodward-Clyde (1990), there are 

seventeen municipally owned and operated pump stations in Santa Clara Valley. These 

pump stations generally consist of pumps, storage units such as a sump or a detention 

basin, and inlet and outlet works. The sumps and detention basins reduce the required 

pump capacity needed to pass the peak flood flows. The pump stations provide flood 

protection to low lying areas that have historically subsided and are now protected by 

levees. The pump stations have generally operated as single-purpose flood control 

facilities. The pumps are designed to go on as soon as water begins to fill the basin 

with the goal of emptying the basin as soon as possible after the event. Therefore, no 

net pollutant removals are assumed to be occurring. 

One retrofitting option to achieve water quality benefits is to change the pump-operating 

schedule in order to increase detention time and to provide for a seasonal wet pond. A 

preliminary evaluation of retrofitting detention basins was encouraging and a pilot study 

to retrofit a facility and perform testing to measure water quality benefits and costs was 

August 1990 through July 1993. 

The following tasks were conducted as part of this study: 

• Retrofit the pump station and modify pump schedules to increase water 

retention and thus improve SW pollutant removal, 

• Conduct water quality sampling to estimate the pollutant removal effectiveness 

of the retrofitted detention basin, and 

• Measure sediment concentrations in the basin in order to determine whether 

sediments are classified as hazardous waste. 

The detention basin has a channel between the inlet and outlet that, prior to the 

modifications, encouraged short-circuiting. A gabion baffle was installed at the outlet to 

reduce short-circuiting and to provide better distribution of flow into the outlet. Rock 

was dumped into the channel leading from the inlet, and a drainage pipe that ran below 

the channel was blocked off. Operational changes consisted of modifying the pump 

schedule to create a 0.61 m (2 ft) permanent pool at the outlet and to provide temporary 

storage and slow release of water over the depth range of 0.61—0.73 m (2—2.4 ft). 

Site Description. The northern portion of Santa Clara Valley has a history of subsidence 

caused by groundwater pumping. In order to protect subsiding areas from flooding, a 

system of levees and pump stations was built. The pump stations collect and pump SW 

runoff from these low-lying areas through the levees into the flood control channels. In 

order to accommodate large flows and reduce the number and pumping capacity of the 

pumps required, some pump stations include relatively large sumps or detention basins. 

An inventory of the pump stations indicated that there are nine such facilities in the 

Valley with relatively large detention basins (Woodward Clyde Consultants 1990). The 

design and operating philosophy of these systems is to: 1) attenuate the peak flow to 

109
 
reduce required pump size, and 2) drain the basins as soon as possible following the 

storm so that flood capacity is available for subsequent storms. 

An example of one of these pump stations is Sunnyvale Pump Station No. 2, located at 

the junction of the Milpitas/Alviso Road (Route 237) and Calabazas Creek. The Pump 

Station consists of four primary pumps, each rated at 1.1 m3

/s (39 cfs) capacity, and 

one auxiliary electric pump with a capacity 0.25 m3/s (9 cfs). The detention basin area is 

approximately 1.8 ha (4.4 ac.) in area and has a capacity of approximately 37,005 m3 

(30 ac-ft) as shown in Figure 5.3.a-1. It receives runoff from a 187.4 ha (463 acre) 

catchment consisting of the following land uses: industrial park (30%), commercial 

(10%), and residential (60%). There is a 2.13 m (7 ft) diameter concrete reinforced pipe 

that drains into the basin. A second 0.91 m (3 ft) diameter line drains a 101.2 ha (250 

acre) catchment (primarily open space) and bypasses the basin to the north and directly 

enters the pump house. 

Treatment Concepts and Retrofitting Objectives. The pump stations could provide an 

opportunity to reduce nonpoint source loads entering the South Bay if they can be costeffectively retrofitted and maintained. The primary treatment is settling of particles. 

Settling can be an effective treatment for some pollutants that are mostly in the 

particulate fraction in stormwater. The particulate fraction of locally collected stormwater 

exhibited a range of 36—94% (mean of 69%) for copper and 24—97% (mean of 66%) 

for lead. These high particulate fractions allow sedimentation to be an effective control 

practice. 

The retrofitting scheme would increase the detention time to allow more particulates to 

settle out into the basin while not significantly increasing the flood risk. A goal of this 

retrofit was to prevent high flows from re-suspending previously settled sediments in the 

detention pond. Scour protection was provided by having at least a two-foot deep 

permanent wet pool during the wet weather season. 

A 24—40 h hydraulic detention time for a pool several feet deep was found to be 

necessary to effectively settle out most of the suspended sediment in the local 

stormwater. 

In all cases, the basin must maintain a relatively large flood control capacity and 

associated outlet works and pumps in order to provide the necessary flood control 

benefits. 

Description of Sunnyvale Retrofit Activities. 

Change Pump Operational Rules to Create a Permanent Pool and Temporary Storage. 

In order to create the permanent pool in the pond, the pumps were set to turn off when 

water levels in the basin (as measured at the outlet) dropped below 0.61 m (2 ft) In 

order to create temporary storage, pump settings were adjusted to phase in (and out) 

very slowly for depths between 0.61—0.73 m (2—2.4 ft). These operational conditions 

created a temporary storage depth above the permanent pool of 0.12 m (0.4 ft) with a 

capacity of 2158.6 m3 (1.75 ac-ft). Because this is an existing flood control facility, the 

110
 
5.3.a-1 Sunnyvale Pump Station No. 2 Showing Retrofit Structures 

111
 
temporary storage depth was determined primarily based on flood control and 

secondarily on water quality considerations. The temporary storage depth was the 

maximum depth that would still allow the basin to pass the 100-year flood. 

Prevent Short Circuiting. The pond contains a trapezoidal open pilot channel (2.44 m (8 

ft) bottom width, 5.2 m (17 ft) top width, and 1.4 m (4.5 ft) depth) between the inlet and 

outlet. In addition to this open channel, a 0.76 m (2.5 ft) reinforced concrete pipe (RCP) 

is situated below the channel to convey low flows between the inlet and outlet. These 

conveyances effectively “short circuited” flows between the inlet and outlet, a condition 

that is highly unsuitable for water quality control. 

In order to limit this short-circuiting, three modifications were made. At the outlet weir 

near the pump house, a gabion baffle was constructed around the original outlet weir to 

prevent short-circuiting of flows along the channel and to promote a better distribution of 

flow from the basin into the outlet weir. A second modification involved placing rocks 

into the channel near the inlet. The third modification involved covering the entrance of 

the 0.76 m (2.5 ft) RCP with a steel plate and vertical riser that reduced the rate at 

which flow would enter the drain below the trapezoidal channel. 

Plug Storm Drain that Directly Entered Pump House. A 0.91 m (3 ft) RCP drained a 

101.2 ha (250 acre) undeveloped area west of the detention basin directly to the pump 

house sump. This pipe was plugged with sandbags in one of the manholes upstream of 

the sump to prevent the runoff from this drainage area from mixing with outflow from the 

detention basin in the sample collection area. 

Problems Encountered. No significant problems were encountered during the structural 

retrofitting of the detention basin. However, the pump control system needed major 

repairs in order to operate the basin within the water level tolerances required for the 

study. Specific problems were encountered with the liquid level sensors and transmitter 

(inaccurate flow monitoring because of the very low flow rates) voltage instabilities 

caused when certain pumps came on line, and fluctuations in the power supply. 

Therefore, an important aspect in evaluating the feasibility of retrofitting pump stations is 

the design and condition of the pump control system and the possible need for repairs 

and upgrading. 

Monitoring Program. The goal of this study was to measure the total runoff and collect 

flow-weighted composite water samples at both the inlet and outlet of the detention 

basin during and after storms in order to estimate pollutant removal performance. 

Sediment samples were also taken to characterize basin sediments. 

Station Design and Equipment. Automated flow and water quality monitoring stations 

were located at the inlet and the outlet of the basin. The inlet pipe was a 2.13 m (7 ft) 

diameter RCP that was quite low, placed at a level below the basin floor, and tended to 

be full of water during most of the wet weather season. The inlet sampling station was 

located 10.7 m (35 ft) upstream of the end of the pipe and consisted of a Druck 

pressure transducer, velocity meter, ISCO Model 3700 automatic water quality sampler 

and Campbell Scientific CR-10 data logger/controller. Flow volumes were estimated 

112
 
using the measured velocity times the wetted area of the pipe and the flow-weighted 

composite samples were to be collected at the inlet based on these estimates. Initially, 

a Montadero-Whitney electromagnetic velocity meter was used. However, the velocities 

in the pipe were too low to measure with this instrument and it was replaced in March 

1992 by a Detectonics I.S. Surveylogger, which relies on the Doppler effect and 

suspended sediment passing the instrument. When compared with estimates of 

anticipated runoff volumes, neither instrument appears to have measured flow velocities 

in a consistent and accurate manner. The primary cause appears to be the relatively 

low velocities in the large pipe. 

The outlet sampling was conducted in the pump house where a Druck pressure 

transducer, an ISCO Model 2700 automatic sampler, and CR-10 datalogger/controller 

were installed. Sampling based on estimated flow through the pumps was planned. 

These estimates were based on the pump run times and pump characteristic curves 

(which show the relationship between flow and head for the design rpm of 700). To 

achieve this, the data logger was connected to the pump house control panel to 

determine pump run times and calculate discharge from the sump. Field visits during 

the 1991—92 season revealed that the pumps did not operate at the design rpm, 

especially during the warm-up period, resulting in inaccurate flow estimates much of the 

time. 

Sampling Methods. At the inlet, a pressure sensor was used to estimate the water level 

in the detention basin. During each sampling event, flow was calculated as a product of 

velocity and area by the CR-10 microprocessor. Based on the flow estimate (which was 

generally poor), the CR-10 initiated and continued water quality sampling at prespecified flow intervals. During a sampling event, instantaneous velocity and pressure 

were recorded each time a water quality sample was taken. Based on anticipated 

rainfall, the sampling algorithm in the CR-10 was designed to instruct the water quality 

sampler to collect twenty 500 ml sub-samples in a 10 liter borosilicate bottle over the 

duration of the storm event. Following the sampling event, the pressure sensor was also 

used to measure water level drops in the pond. 

At the outlet, average hourly flowrate was estimated based on the pump run times and 

the pump characteristic curves (also inaccurate), and was recorded over the duration of 

the wet-weather season. To begin an event, field crews manually initiated the 

automated samplers based on anticipated flow volumes for that storm. As with the inlet, 

the automated samplers recorded instantaneous flow measurements when each 

sample was collected. 

Stations were visited prior to, during, and after monitored events to ice samples, 

exchange sample bottles, and ensure proper equipment operation. Conductivity, pH, 

and temperature were measured during the site visits. 

Data Collected. Eight storm events were sampled. For six of these events, flowweighted composite water quality samples and hydrologic measurements were taken at 

both inlet and outlet stations. In most cases, only partial flow measurements were made 

113
 
because equipment malfunctioned, threshold velocities were not achieved, or there 

were problems with the pump control system. 

Due to the uncertainty in flow volume measurements, pollutant loads were not used to 

estimate treatment effectiveness. Instead, effectiveness was estimated based on the 

flow composite water quality concentration data, assuming that the inlet and outlet 

volumes for an event are equal. 

Sediment samples were taken at three locations: in the center of the basin, near the 

inlet, and near the outlet. Three sets of sediment samples were collected during dry 

periods when the basin was empty, or nearly empty, of water (June 15, 1990, May 14, 

1992, and on July 12, 1993). The first samples were obtained using a 10.2 cm (4 in) 

stainless steel hand auger, while the other samples were collected by scraping the top 

1.27 cm (0.5 in) of sediment with a Teflon� -lined scraper. 

Flooding Analysis, Storm Hydrology, Water Quality and Sediment Monitoring Results 

Flooding Analysis. Woodward Clyde used a reservoir routing model to estimate water 

levels in the basin for the 100-year inflow event and for two pump operating scenarios. 

The first pumping scenario corresponded to the original pumping schedule used for 

flood control. The second scenario corresponded to the revised pumping schedule 

appropriate for a multipurpose flood control and water quality control facility. Based on 

the results of the model, the maximum water level in the basin for the 100-year flood did 

not change by modifying the pump operation schedule. 

Precipitation. Rainfall was measured with a tipping bucket, which registered the time 

when the bucket collected 0.25 cm (0.1 in) of rainfall. The ranges of storm volumes 

during the sampling period were from 1—5.6 cm (0.4—2.2 in) and the storm durations 

ranged from 6—60 h. The Synoptic Rainfall Analysis Program (SYNOP) examined the 

long period characteristics of the local historical rainfall data that was collected by the 

NWS at the San Jose Airport (Gage No. 7821). The median event rainfall volume for the 

San Jose Airport gage for the period from 1948—1989 was 1.27 cm (0.5 in). 

Runoff. Flow measurements collected at the inlet and outlet for various events were 

compared with rainfall to calculate the volumetric runoff coefficients. The flow 

measurements at both the inlet and outlet stations were not considered very reliable, as 

the measured runoff coefficients ranged from 0.1—1.89. Woodward Clyde estimated 

that the actual values would be about 0.5—0.8 for these rains and watershed 

characteristics. 

Comparison of Inlet Water Quality to Other Santa Clara SW Monitoring Station Data. 

Laboratory chemical analyses were conducted on the water samples collected at the 

basin inlet and outlet stations during the six storm events. The median flow-weighted 

composite concentrations of total metals (cadmium, chromium, copper, lead, nickel, and 

zinc) from the inlet station are summarized in Table 5.3.a-1. The table shows median 

concentrations obtained from other Santa Clara Valley SW monitoring stations 

representing residential-commercial, industrial, and open land uses. The inlet 

concentrations of copper, lead, nickel, and zinc are higher than concentrations from 

114
 
open land use, but lower than concentrations at residential-commercial and industrial 

land use stations. The cadmium concentration appears to be very similar to the 

residential-commercial land use, while the chromium concentration is closer to the open 

land use. 

Table 5.3.a-1. Comparison of Median Metal Concentrations at Inlet to Retrofitted Basin 

to other Santa Clara Valley Stormwater Monitoring Station Data 

Inlet to Retrofit Residential/Commercia Industrial Land Open Space Land 

Metal Basin l Land Use Station Use Station Use Station 

(n=6) (n=21) (n=25) (n=4) 

(mg/l) (mg/l) (mg/l) (mg/l) 

Cadmium 1.1 1.0 3.9 0.3 

Chromium 12 16 24 11 

Copper 24 33 51 11 

Lead 38 45 91 2.0 

Nickel 21 30 46 5.0 

Zinc 180 240 1,150 5.0 

*n= number of storm events 

Pollutant Removal Effectiveness. Table 5.3.a-2 summarizes inlet and outlet 

concentrations for total and dissolved metals (cadmium, chromium, copper, lead, nickel, 

and zinc), SS, hardness and total oil and grease for several storm events. Based on 

these data, pollutant reductions were estimated as the outlet minus inlet concentration 

divided by the inlet concentration. The average pollutant removal effectiveness for the 

metals ranged from about 30—50%. The metals removal data indicated that the 

removal of total chromium, copper, lead, nickel and zinc were well correlated with SS 

removal. 

Comparison to Water Quality Objectives (WQOs). Of these metals, total and dissolved 

chromium, lead and nickel did not exceed the acute WQOs. Total and dissolved 

cadmium exceeded the WQO in only one storm out of six monitored storms. Total 

copper at the inlet station exceeded WQOs in four out of six storm events. However, 

concentrations at the outlet station never exceeded WQOs (though the outlet 

concentration was essentially equal to the WQO for one event). None of the dissolved 

copper concentrations exceeded the acute WQOs. Total zinc concentrations at the inlet 

and outlet stations exceeded the acute WQOs for all six storms. Dissolved 

concentrations of zinc at the outlet station exceeded the WQOs in three of the six 

events. 

Sediment Quality. The objectives of the sediment sampling was to characterize 

sediment quality in the detention basin and to compare the sediment concentrations to 

hazardous waste criteria. Results of these sediment samples are summarized in Table 

5.3.a-3. 

115
 
-- -- -- -- -- -- -- --

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-- --

-- --

-- -- -- -- -- -- -- -- --

-- -- -- -- -- -- -- -- --

-- -- -- -- -- -- -- -- --

-- -- -- -- -- -- -- -- --

-- -- -- -- -- -- -- -- --

Table 5.3.a-2. Inlet and Outlet Observed Concentrations and Pollutant Removals 

Cadmium 

(mg/L) 

Chromium 

(mg/L) 

Copper 

(mg/L) 

Lead 

(mg/L) 

Nickel 

(mg/L) 

Zinc 

(mg/L) 

SS 

(mg/L) 

TH 

(mg/L) 

O&G 

nf f nf f Nf f nf f nf f nf f 

SE 17 

Inlet 0.4 <0.2 3.6 1.8 8.7 5. 6.4 2. 1.7 <2 46 28 12 97 1.5 

4 2 

Outlet 0.2 <0.2 2.7 1.1 6.8 4. 3.4 1 1.7 <2 26 19 73 120 1.4 

7 

Reduction 25% 22% 47% 0% 43% 39% 7% 

SE 20 

Inlet 6.6 1.3 12 1 24 3 45 1 16 1 180 19 90 110 0.2 

Outlet 4.8 2.5 6 1 9 3 10 1 4 1 73 22 24 63 <0.2 

Reduction 50% 63% 78% 75% 59% 73% 

SE 21b 

Inlet 1.1 0.2 18 1 24 2 53 <1 25 <1 180 5 140 

Outlet 1.5 <0.2 14 1 16 2 35 <1 19 <1 120 7 93 

Reduction 22% 33% 34% 24% 33% 34% 

SE 23 

Inlet 1 0.2 11 <1 27 5 30 1 13 3. 190 41 74 100 0.7 

9 

Outlet 0.6 <0.2 8.3 1.4 12 4. 12 <1 5.8 2. 82 45 31 90 0.5 

7 2 

Reduction 25% 56% 60% 55% 57% 58% 

SE 24 

Inlet 1.6 <0.2 21 1.1 40 2. 76 <1 42 9. 270 22 180 140 0.6 

1 6 

Outlet 1.3 0.2 15 8.6 24 5 40 1. 29 15 160 31 96 140 3.5 

4 

Reduction 29% 40% 47% 31% 41% 47% 

SE 27 

Inlet 1 0.5 6.3 1.4 14 5. 13 <1 83 63 70 35 30 110 1.6 

4 

Outlet 0.6 0.4 4.9 1.7 8.9 4. 6.6 <1 25 20 47 26 15 220 1.3 

5 

Reduction 22% 36% 49% 70% 33% 50% 

Average 

Reduction 29% 42% 53% 51% 44% 50% 

nf: non-filtered (total) SS: suspended solids O&G: oil and grease 

f: filtered (“dissolved”) TH: total hardness, as CaCO3 

Removals are only given if most observations were >PQL 

116
 
-- -- -- --

-- --

Table 5.3.a-3. Sediment Observations (mg/kg) 

% TOC Cadmium Chromium Copper Iron Lead Manganese Nickel Zinc 

6/15/90 2.2 92 36 61 320 

core 

5/14/92 

surface 3.8 23 200 150 49,000 280 610 94 750 

Inlet 

Middle 5.5 17 220 140 38,600 350 640 87 570 

Outlet 1.9 35 140 47 47,700 18 680 76 260 

7/12/93 

surface surface 2.4 1.0 170 110 34,000 260 560 96 220 

Inlet 

Middle 0.65 0.2 120 37 36,000 12 700 75 85 

Outlet 0.93 0.3 110 43 30,000 24 570 73 63 

TTLC100 2,500 2,500 1,000 2,000 5,00 

0 

TTLC: Total Threshold Limit Concentration 

In the second and third rounds of sampling, the highest concentrations for copper, 

nickel and zinc were found at the inlet station. Cadmium, chromium and lead were also 

highest at the inlet station for the July 12, 1993 sampling round. The high concentration 

of the majority of the metals near the inlet is consistent with other studies. 

Average sediment concentrations observed in Pump Station No. 2 are compared in 

Table 5.3.a-4 with sediment data collected from other detention basins in the Valley and 

elsewhere. Results from the various basins differ substantially and indicate that 

sediment quality is highly site specific and varies depending on soils, catchment land 

use, and other factors, especially time when the samples were analyzed (for lead). 

To evaluate whether the sediments were hazardous, concentrations were compared to 

standards established in the California Administrative Code, Title 22. Under Title 22, 

there are two criteria for designating solids as hazardous waste. The first criterion is that 

the sediment concentrations not exceed the Total Threshold Limit Concentrations 

(TTLC). The second criterion is that the extract obtained from the whole effluent toxicity 

(WET) extraction method not exceed the Soluble Threshold Limit Concentrations 

(STLC). For this pilot scale screening level of analysis, it was considered adequate to 

compare with the TTLC only. In situations where disposal is being considered, the WET 

extraction test should also be conducted. 

None of the sediment sample concentrations collected in the Sunnyvale Pump Station 

basin exceeded the TTLC. The highest concentrations of cadmium, lead, and zinc were 

4, 3, and 7 times lower than the TTLC, respectively. The highest concentrations 

reported for chromium, copper, and nickel were 11, 17 and 21 times lower than the 

TTLC, respectively. Based on these concentrations, sediments are not considered 

hazardous. 

117
 
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-- --

-- --

-- --

--

--

-- --

-- --

--

--

--

Table 5.3.a-4. Comparison of Average Sediment Concentrations from Detention Basins 

and Swales (mg/kg) 

Detention Basin Cadmium Copper Nickel Lead Zinc 

This Retrofit Basin 11.2 88 80 140 324 

Other Santa Clara County
 

Eastside Basin A
 0.37 32 36 17 68 

Eastside Basin B
 0.37 36 40 6 73 

Eastside Basin C
 1 71 100 11 330 

River Oaks
 Nd 24 72 14 84 

Fresno NURP
 

Recharge F
 37 32 713 

Recharge G
 25 37 487 

Recharge M
 55 53 1333 

Recharge EE
 25 22 297 

Recharge MM
 9.5 11 93 

Wigington (1983)
 

Bulk Mail Basin
 2.8 19 112 224 

Kmart Basin
 0.8 13 368 114 

Nightingale (1975)
 

Detention Basin
 20 224 107 

Special Pit
 23 801 236 

Wigington (1986)
 

Fairidge Swales
 0.26 4.2 42 102 

Stratton Woods Swales
 0.18 10 18 70 

Rte. 234 Rd. Swales
 0.82 23 936 106 

Cost Effectiveness Evaluation 

The mean annual runoff volume of 433,000 m3

 (351 ac-ft) was estimated based on 

mean annual rainfall of 33 cm (13 in.) in the vicinity of the basin, an assumed runoff 

coefficient of 0.7, and the 187.4 ha (463 ac.) catchment area. Mean concentrations and 

removal efficiencies are averages of observed data. For the metals, annual load 

reductions ranged from 272 grams (0.6 lbs) for cadmium to 29.5 kg (65 lbs) for zinc. For 

copper, the annual load reduction is estimated at 4.1 kg (9 lbs), which represents 

approximately 40% of the total copper that enters the basin. Table 5.3.a-5 summarizes 

the estimated cost effectiveness for the removal of heavy metals from the pond. 

118
 
Table 5.3.a-5. Estimated Mean Annual Load reduction and Cost Effectiveness* 

METAL Mean Inlet Average Annual Load Cost 

Concentration Removal Reduction Effectiveness 

(mg/L) Efficiency (%) (kg/yr.) (kg/$1,000) 

Cadmium 0.002 35 0.3 0.03 

Chromium 0.012 29 1.5 0.18 

Copper 0.023 42 4.2 0.5 

Lead 0.037 53 8.2 1.0 

Nickel 0.038 51 8.2 1.0 

Zinc 0.156 0.44 29.5 3.6 

SS 87 0.50 18,600 2,270 

*Assuming an annual runoff volume of 433,000 m3

 (351 ac-ft) 

Solids Accumulation and Removal. About 18,600 kg (41,000 lbs) of SS would be 

collected annually in the retrofitted detention basin, which represents about one-half of 

the annual input of solids. Assuming a specific gravity of about 1.5 for saturated 

sediment collected in the basin, this would correspond to about 12.2 m3 (16 cyd) of 

material annually. If uniformly distributed over the 1.78 ha (4.4 acre) basin, the mean 

annual accumulation rate would be 0.08 cm per year (0.03 in. per year). Sediments are 

expected to accumulate near the inlet and, in this specific case, in the pilot channel. In 

ten years, this accumulation rate would equal about 123.3 m3 (0.1 ac-ft) compared with 

the capacity of the basin that is 37,000 m3 (30 ac-ft). Therefore, this accumulation of 

sediments does not pose a risk to reducing the flood control capacity of the basin. 

Accumulation of at least 15.24 cm (6 in.) of sediment is required before removal is 

practical. This amount of sediment may take as long as 10 or 20 years to accumulate. 

Capital, Operation and Maintenance Costs. Capital and O&M costs were estimated for 

the retrofitted pump station and are shown in Table 5.3.a-6. Costs were classified as 

capital expenditures, O&M, and disposal. Capital costs for the structural retrofitting were 

based on actual costs; whereas the costs for repair of the pump electronic control 

systems were estimated. Operations and maintenance assumes 100 hours per year 

labor in addition to that already being conducted to operate and maintain the facility for 

flood control. Disposal costs assume disposal is conducted every 10 years and include 

estimated future costs for landfill fees, trucking, and excavation. The total annualized 

cost is therefore estimated to be $8,200 for the 187.4 ha (463 acre) watershed, or about 

$7.3 per ha ($18 per acre) of watershed per year. The removal costs for copper were 

estimated to be about 0.5 kg per $1,000 (1.1 lbs per $1,000) that compares very 

favorably with other stormwater control alternatives, e.g. copper removal via street 

cleaning is approximately 0.7 kg per $1,000 (1.5 lbs per $1,000) (Pitt 1979, 1985, and 

1987). 

119
 
Table 5.3.a-6. Estimated Annualized Costs for Capital Expenditures and Operation 

CATEGORY COST 

1. Capital Expenditures 

Structural retrofitting = $15,000 

Amortized over 20 years at 8% $1,500 

2. Operations and Maintenance 

Inspection and repair (100 h @ $50/h) $5,000 

3. Disposal (every 10 years) 

Landfill (122 m3 @ $38/m3 $8,000 

Trucking (16 trips @$75/h x 2 h/trip $2,400 

Excavation (122 cyd @ $7.7/m3 $1,600 

Amortized over 10 years at 8% $1,700 

4. Total Cost per Year $20,200 

Conclusions 

Implications for Other Facilities. According to an inventory conducted by WoodwardClyde in 1990, nine of the existing 17 municipal pump stations in Santa Clara Valley are 

designed with detention basins (rather than sumps) and are, therefore, suitable for 

comparison with the pilot project. The detention basins range from 0.6—5.7 ha (1.5—14 

ac.), with capacities of 5,550—182,556 m3

 (4.5—148 ac-ft). The watershed area for 

each pump station ranges from 10.1—405 ha (25—1,000 ac.) and the total watershed 

area served by all nine stations is 1,725 ha (4,260 ac.) or 17 km2 (6.6 mi2

). This is about 

2% of the 900 km2 (350 mi2) area of the Santa Clara Valley below the upland reservoirs. 

If we assume that other similar facilities could be retrofitted to achieve a performance 

comparable to that measured at Pump Station No. 2, the net reduction in copper load to 

the Bay would be about 45 kg (100 lbs). This is only about 1% of the estimated mean 

annual load of 6,350 kg (14,000 lbs) of copper entering San Francisco Bay. 

•	 A 100-year flood analysis indicated that modification of the pump schedule to 

achieve water quality benefits did not increase the maximum 100-year elevation in 

the pond. 

•	 Based on measured inlet and outlet flow composite concentrations from 6 storm 

events, the average pollutant removal efficiencies were: total chromium, 29%; total 

copper, 42%; total lead, 53%; total nickel, 51%; total zinc, 44%; and total SS, 50%. 

• The removal efficiencies for chromium, copper, lead, nickel and zinc correlated 

well with SS removal, indicating that SS may be used as a surrogate parameter to 

monitor effectiveness of metals removal in detention basins. 

•	 Metal concentrations of basin sediments were generally highest at the inlet 

location. 

120
 
•	 None of the sediment concentrations exceeded the TTLC standard, indicating that 

the sediments are not hazardous. 

• The estimated mean annual load reduction of metals ranged between 2.7—29.5 

kg (0.6—65 lbs), depending on the metal. The mean annual load reduction for 

copper was 4.1 kg (9 lbs) 

•	 The amortized annual capital and O&M cost for retrofitting the Sunnyvale Pump 

Station No. 2 is estimated at $8,200. The cost-effectiveness removal rate for 

copper is 0.5 kg/$1,000 (1.1 lb/$1,000). 

•	 Solids accumulation rates are very low and are estimated to be approximately 123.3 

m3

 (0.1 ac-ft) over 10 years. Given that the basin has a capacity of 37,000 m3 (30 

ac-ft), increased deposition caused by retrofitting does not increase flood risk. 

Implications for Management. 

•	 The total watershed area in Santa Clara Valley served by the nine pump stations 

with retention basins is approximately or 17 km2

 (6.6 mi2) (only 2% of the total area 

of the Valley downstream of the reservoirs). Thus, even if an improved treatment 

performance could be obtained from these basins, the total load reduction to the Bay 

would be minimal. For example, the load reduction of copper would only be 45.4 kg 

(100 lbs), which is less than 1% of the estimated mean copper load to the Bay. 

•	 Since pump stations are relatively easy to retrofit, water quality benefits could be 

achieved by simply changing the pumping schedule. 

•	 If a retrofitting program is to be pursued, it would be important to ensure that the 

pump control equipment is operational and well maintained, and that staff are well 

trained in its use. 

5.3.b. Monroe Street Detention Pond, Madison, Wisconsin 

Introduction and Site Description 

This case study considers the benefits of retrofitting the outlet works of an existing wet 

detention pond. The original pond was creating severe downstream erosion in the 

channels between the pond and the receiving water, and the pond storage volume was 

not effectively being used for either flood control or water quality benefits. The outlets 

were modified and the pond has undergone extensive monitoring to confirm the water 

quality benefits of the retrofit. 

The US Geological Survey (USGS), in cooperation with the Wisconsin Department of 

Natural Resources (WDNR) investigated the Monroe St. wet detention pond located in 

Madison, WI (House, et al. 1993). The University of Wisconsin originally constructed the 

pond to protect the water quality and ecology of Lake Wingra and surrounding wetlands 

from stormwater. The pond is located on the downstream side of Monroe Street at the 

121
 
outlet of a storm sewer that drains a 0.96 km2

 (237 acre) urbanized area. Land use in 

the watershed area consists mostly of single-family residences and commercial strip 

development, with some institutional uses (i.e., schools and churches). The average 

basin slope is 2.2%. 

The Monroe Street pond has a surface area of 5,670 m2

 (1.42 acre), a maximum depth 

of 2.3 m (7.5 ft) and an average depth of 1.1 m (3.6 ft) at normal pool elevation. The 

shape of the pond is round to oval with a small island. The inlet side is nearest to 

Monroe Street and the two outlets are on the far side away from Monroe Street. The 

pond has a surcharge storage volume above the normal pool elevation that is capable 

of holding the 10-year, 24 h storm-runoff volume without overtopping the containment 

berm around the pond. The pond has two outlets, each controlled by 90� V-notch weirs 

that drain to channels leading to Lake Wingra. The weirs are located in 2.4 m (8 ft) 

diameter concrete vaults, with 0.8 m (2.5 ft) concrete pipes leading to the pond. The 

initial primary outlet configuration consisted of two 2.4 m (8 ft) long rectangular weirs 

located in the vaults, made with concrete block walls. The original flow capacity of these 

two weirs was enormous, being about 1.4 m3/s (50 cfs) at 0.3 m (1 ft) head and 7 m3/s 

(250 cfs) at 0.9 m (3 ft) of head. The outlets in the pond are submerged and the bottom 

of the pond consists of a clay layer that inhibits infiltration of water from or into the pond. 

As noted above, the discharges from the pond were little attenuated from the inflow 

velocities and severe channel erosion was occurring in the wetlands, negating the 

sediment trapping benefits of the pond. The emergency spillway appeared to never 

have been used since construction, even with several massive storms. In fact, the pond 

elevation barely fluctuated. The water quality benefits of the existing dry pond were 

assumed to be negligible. 

The outlets were therefore modified in 1995 to reduce the downstream erosion 

problems by removing several courses of concrete blocks and installing 90� V-notch 

weirs made of plate steel in each vault. The pond normal water level was dropped 

about 15.3 cm (6 in.) with a lowered invert. The new primary outlets have total flow 

capacities of about 0.14 m3

/s (5 cfs) at 0.3 m (1 ft) head and 2.3 m3/s (80 cfs) at 0.9 m 

(3 ft) head. The pond surface fluctuates more now, and the emergency spillway has 

been active every few years. Most significantly, the downstream channels are now 

stable. 

The pond was designed for an expected 90% event mean concentration (EMC) removal 

for SS (particulate residue). The ratio of pond to drainage area is 0.6%. This 

percentage is close to the value (0.4—0.8%) required for 5 micron control, which 

generally corresponds to a 90% reduction of SS. 

The Monroe Street pond may have more water quality monitoring data than any pond in 

the country. A total of 64 events were extensively monitored between February 1987 

and April 1988. The monitored rains varied from 2 to more than 82 mm during this 

period. Periodic water quality and flow monitoring has also continued at this pond since 

1988. 

122
 
Method of Investigation 

Water-quality data were collected by the U.S. Geological Survey (House, et al. 1993) 

using programmable automatic refrigerated water samplers installed at the inflow and 

outflow sites of the pond. The outflow data was collected at two locations, east and 

west. The samplers were programmed to obtain flow-proportional samples for each 

storm. These samples represent the flow-averaged constituent concentrations during a 

runoff event. These samples were removed from the samplers, preserved, and shipped 

to the Denver USGS laboratory for analysis within 24 h of being collected. The samples 

were analyzed for SS, volatile total solids, total and dissolved chemical oxygen demand 

(COD), total chloride, total and dissolved phosphorus, phosphate total and dissolved 

forms of TKN, nitrates, and total and dissolved forms of copper, zinc, and lead. 

Precipitation data were also recorded at 5 min intervals during the storm events using a 

recording rain gage located at the pond site. Storm runoff (pond inflow) was monitored 

at the box culvert that was the terminus of the 0.96 km2 drainage area. Discharge rates 

and flow volumes passing through the culvert were determined by use of a flow velocity 

sensor and water level indicator installed inside the culvert. The velocity and depth 

sensors were connected to a data logger that recorded the water level and velocity data 

and computed discharge rates based on the culvert geometry. 

Data Analysis and Observations 

The pond inlet and outlet pollutant concentrations were analyzed to determine the 

pollutant reduction. Various relationships between inlet and outlet concentrations were 

investigated by statistical analysis, including Particulate Pollutant Strength (PPS) and 

percent controls. Each statistical process is described in the following paragraphs. The 

basic data are contained in the USGS report. (House, et al. 1993). 

Hydrograph/Flow Calibration. An important part of the Monroe Street project was 

validating the DETPOND wet detention pond water quality model that was used to 

design the retrofit of the outlet structures (Pitt and Voorhees 1995). The first step in the 

validation was to check flow volumes and peak flow rates, and the complete 

hydrographs. 

Fifteen storm events were used to validate the flow portions of the DETPOND program. 

The program predicted outflow flow values from the inflow hydrographs using the 

modified pulse routing method. The outfall predictions (at five-minute intervals) were 

compared to the observed outfall flow values. The predicted outflow hydrographs very 

closely matched the corresponding observed outflow hydrographs. In addition to 

comparing the general shape of the discharge hydrographs, the outflow total discharge 

volume, peak discharge flow rate, SS removal, and outflow particle size distribution 

were also compared for validation. The predicted outflow volumes and peak discharges 

very closely matched the observed outflow conditions. 

Observed Influent and Effluent Pollutant Concentrations. Table 5.3.b-1 lists the influent 

and effluent conditions observed at the pond. The number of observations, and the 

mean, maximum, and average concentrations are shown. The standard deviation (Std. 

123
 
Dev.) and coefficient of variation (COV) are given. The Mann Whitney a values, which 

indicates whether the influent and effluent values are statistically significantly different, 

are also summarized. 

Table 5.3.b-1. Summary of Observed Influent and Effluent Pollutant Concentrations at 

Monroe St. Pond 

RESIDUE 

Number of 

Observations Minimum Max. Mean 

Std. 

Dev. COV 

* 

Mann 

Whitney 

Total Solid (mg/L) 

IN 64 76 5810 526 816 1.55 0.0003 

OUT 57 63 1180 276 259 0.94 

Suspended Solid (mg/L) 

IN 64 21 1330 230 257 1.11 0.0000 

OUT 57 4 140 32 28 0.87 

Filtered Residue (mg/L) 

IN 64 303 789 2.6 0.0758 

OUT 57 244 244 1.0 

Volatile Residue (mg/L) 

IN 62 28 376 92 63 0.68 0.0000 

OUT 55 0 232 48 30 0.63 

Total Chlorides 

Number of 

Observations 

Minimum Max. Mean Std. 

Dev. 

COV Mann 

Whitney 

IN 65 0.8 3100 120 430 3.58 .0032 

OUT 57 2.1 570 90.9 135.5 1.49 

COD 

Total COD, (mg/L) 

IN 63 21 370 90 63.4 0.70 0.000 

OUT 57 16 350 37 44 1.2 

Particulate COD, (mg/L) 

IN 56 8 327 53 53 1.0 0.000 

OUT 52 -3 281 16 38.5 2.4 

Filtered COD, (mg/L) 

IN 56 10 160 38 32.3 0.85 0.0019 

OUT 52 9 69 21.4 11 0.51 

* Coefficient of Variation – standard deviation divided by the mean. 

124 
Table 5.3.b-1. Summary of Observed Influent and Effluent Pollutant Concentrations at 

Monroe St. Pond, Continued 

RESIDUE 

Number of 

Observations Minimum Max. Mean 

Std. 

Dev. COV 

Mann 

Whitney 

PHOSPHORUS 

Total Phosphorus, (mg/L) 

IN 65 0.16 1.7 0.56 0.39 0.69 0.000 

OUT 59 0.13 1.6 0.26 0.19 0.73 

Particulate Phosphorus, (mg/L) 

IN 60 0.08 1.53 0.33 0.27 0.82 0.000 

OUT 53 0.04 1.52 0.16 0.2 1.25 

Filtered Phosphorus, (mg/L) 

IN 60 0.06 1.33 0.23 0.28 1.2 0.0001 

OUT 53 0.02 0.27 0.1 0.06 0.6 

PHOSPHATES 

Total Phosphates, (mg/L) 

IN 56 0.026 1.22 0.2 0.27 1.35 0.0002 

OUT 51 0.004 0.25 0.08 0.06 0.75 

COPPER 

Total Copper, (mg/L) 

IN 64 10 130 50 14 0.28 0.592 

OUT 57 3.5 60 46 12.43 0.27 

Particulate Copper, (ug/L) 

IN 60 7 120 43 13.93 0.32 0.0652 

OUT 52 .5 49 42.3 12.02 0.28 

Filtered Copper, (ug/L) 

IN 60 3 16 6.35 3.32 0.52 0.0035 

OUT 53 3 13.5 4.5 2.17 0.48 

LEAD 

Total Lead, (ug/L) 

IN 64 18 420 85 52.4 0.61 0.100 

OUT 57 4 100 68 39 0.57 

Particulate Lead, (ug/L) 

IN 59 15 417 83 54 0.65 0.1800 

OUT 52 1 97 66 39 0.59 

125
 
Table 5.3.b-1. Summary of Observed Influent and Effluent Pollutant Concentrations at 

Monroe St. Pond, Continued 

RESIDUE 

Number of 

Observations Minimum Max. Mean 

Std. 

Dev. COV 

Mann 

Whitney 

Filtered Lead, (ug/L) 

IN 59 3 12 3.5 1.7 0.49 0.54 

OUT 53 3 4 3.0 0.17 0.06 

ZINC 

Total Zinc, (ug/L) 

IN 64 35 1100 152 136 0.89 0.0450 

OUT 57 30 1400 191 265 1.39 

Particulate Zinc, (ug/L) 

IN 58 -3.0 980 103.5 131 1.26 0.0000 

OUT 47 -15.5 630 60.3 101 1.68 

Filtered Zinc, (ug/L) 

IN 57 9.0 160.0 51 34 0.66 0.0020 

OUT 48 20.0 1218 150 215 1.43 

The coefficient of variation (COV) is calculated by dividing the standard deviation by the 

mean and normalizes the standard deviation. A high value indicates that the data 

spread is wide, requiring many data observations to obtain a precise estimate of the 

EMC. 

COV varied from 3.58 for inlet chlorides, which may not be significant due to seasonal 

variations, to 0.06 for outlet filtered lead. All filtered lead had lower COV values which 

indicated small concentration variations for samples. 

The Mann-Whitney test is a non-parametric analysis comparing two sets of data. The 

null hypothesis used in the Mann-Whitney is inlet pollutant concentration minus outlet 

pollutant concentration equals zero. Generally, an a value < 0.05 indicates that the 

sample sets are from two different populations (significantly different at the 95% 

confidence level). The following constituents had significant a values indicating that the 

pond significantly affected the concentrations: 

Total solids Phosphate
 

Suspended solids TKN (total and particulate)
 

Volatile solids Nitrate
 

Chlorides Copper (filtered)
 

COD (all forms) Zinc (all forms)
 

Phosphorus (all forms)
 

126
 
Particulate Pollutant Strength. PPS is the ratio of a particulate pollutant concentration 

to the suspended solid concentration expressed in mg/kg. PPS was calculated for each 

pollutant with a particulate form. The small particles in stormwater have greater PPS 

values than the large particles. Therefore, all pollutants had higher outlet than inlet PPS 

values due to preferential removals of large particles in the detention pond, leaving 

relatively more small particles in the discharge water. A wide difference between influent 

and effluent PPS values indicates that the main components of the contaminant (such 

as for TKN and phosphorus) are associated with fine particles. 

Control of Pollutants. The reduction of pollutants was calculated from the difference in 

pollutant concentrations in the inlet and outlet water for each event, as shown on Table 

5.3.b-2. As expected, control was higher for all particulate forms of the constituents 

than for filtered forms, except for copper. Filtered constituents (<0.45 mm) have 

characteristics like colloidal particles and will tend to be transported through the wet 

detention pond relatively unchanged. A well designed wet detention pond will remove 

70—90% of SS, 70% of COD, 60—70% of nutrients and 60—95% of the particulate 

forms of the heavy metals. 

Table 5.3.b-2. Summary Table of Pollutant Control* 

Parameter 10% 50% 90% 

Suspended solids 35 87 97 

Total Residue <0 52 86 

Volatile Residue <0 41 76 

Filtered Residue <0 <0 56 

Particulate COD 15 80 95 

Total COD 29 60 84 

Filtered COD <0 24 80 

Particulate -20 60 80 

Phosphorus 

Total Phosphorus <0 47 81 

Filtered Phosphorus <0 43 83 

Particulate TKN -40 40 80 

Total TKN <0 45 75 

Filtered TKN <0 12 68 

Particulate Zinc - 117	 70 95 

Total Zinc <0 31 69 

Filtered Zinc <0 <0 59 

* Copper and Lead observations were mostly below the detection limits 

and are therefore not shown. 

127
 
Particle Size Distributions and Short-Circuiting 

Seven events were studied to find the short-circuiting “n” factors using observed and 

predicted particle size distributions in effluent water. Particle size distributions were 

measured using the Sedigraph method at the USGS Denver laboratory. This technique 

measures settling rates of different size suspended solid particulates down to 2 microns. 

The value of n is calculated using the concentrations of large particles that are found in 

the effluent. In ideal settling, no particles greater than the theoretical critical size (about 

5 mm for Monroe Street) should appear in the effluent. However, there is always a small 

number of these larger particles in the effluent. It is generally assumed that shortcircuiting is responsible for these large particles. The measured values for n were one, 

or less, indicating a high degree of short-circuiting in the pond. However, these 

observations were possibly affected by scour of bottom deposits near the subsurface 

effluent pipes. The maximum effect of short-circuiting on pond performance is shown in 

Table 5.3.b-3, showing the average reduction in SS removals for different n values, 

compared to the best performance (n value equal to 8): 

Table 5.3.b-3. Average Suspended Solids Removal for variable n values for the 

Monroe 

St. Ponds 

n value % SS removal Reduction in % SS 

(average) removal compared to n=8 

8 85 

3 84 1 

1 80.7 4.3 

0.5 78.5 6.5 

0.2 59 26 

The median value of n observed was about 0.35, indicating a reduction in SS capture 

efficiency of no more than about 10%. The effects of this short-circuiting, has a minimal 

effect on the SS percentage removals. The Monroe Street pond provided an average 

SS reduction of 87%, compared to the design goal of 90%. The small difference (3%) 

between actual and design SS reduction indicates that the short-circuiting has a 

negligible effect on actual performance. 

Monroe St. Pond Retrofitting Conclusions 

The retrofitting of the Monroe Street wet detention pond was very successful. Changing 

the outlet structures from large rectangular weirs to V-notch weirs significantly reduced 

effluent flows, reduced downstream channel erosion and improved water quality 

benefits of the pond. 

All effluent concentrations were lower than associated influent concentrations, except 

for chlorides. The measured short-circuiting factor indicated a severe short-circuiting 

problem, but that could be a false indication due to minor scour near the effluent works 

128
 
in the pond. Data suggest that short circuiting does not adversely impact treatment 

efficiency. The Monroe Street pond is meeting all reasonable expectations in both 

downstream channel protection and in contaminant capture. 

5.3.c. Birmingham, Alabama Dry-Ponds 

This retrofit case study examines a dry detention pond at a new apartment complex in 

Birmingham, Alabama. This is an example of a typical small pond installed to meet local 

regulations for peak flow control. The pond is not even expected to operate satisfactory 

even as a peak flow rate pond based on review of the current hydrological assessment. 

This analysis makes suggestions for retrofitting this pond to a wet pond and to improve 

the overall benefits of the device. It also demonstrates how a combination of water 

quality and flow analysis computer programs can be used together to evaluate and 

design effective multiple benefit SW controls. The water quality benefits of the existing 

dry pond were assumed to be negligible for this analysis. 

Redesign of Dry Pond for Water Quality Benefits 

There are a multitude of options for retrofitting the existing dry pond: excavation below 

the outlet, raising the outlet, or some combination of the two. However, since relocating 

and/or reconstructing the existing outlet structures to gain additional water volume 

below the outlet would be quite expensive we will only consider excavating below the 

existing outlet located at 219.5 m (720 ft) mean sea level (msl). This redesign will only 

require excavation within the existing space with a minimum of re-grading. 

There are several factors to consider when designing a wet pond, among these are 

depth, safety, and storage criteria. Pitt (1998) recommends 0.9—1.8 m (3—6 ft) of 

storage below the inve rt of the lowest outlet, 219.5 m (720 ft). In addition, safety 

requires a shallow ledge, submerged 0.3 m (1 ft) below the permanent storage area. 

The redesign presented here will include the preferred depth guideline of 1.8 m (6 ft), 

requiring excavation from 219.5—217.6 m (720—714 ft), and a minimal three-foot ledge 

at 219.2 m (719 ft) to match existing grades. The side slope will be 1:1 from 219.5 m 

(720 ft) to the ledge at 219.2 m (719 ft), and also 1:1 from the interior of the ledge to 

217.6 m (714 ft). Assuming a prismatic cross-section, the additional wet storage to be 

constructed below the 219.5 m (720 ft) of elevation is approximately 250 m3

 (0.204 acft). 

Safety Criteria. Pitt (1998) recommends a gradual slope, 1:1 to 1:4, on the approach to 

the pond’s normal high water mark, 221.6 m (727 ft) in this case. The existing pond 

may be dangerous, as the side slope is approximately 1:2 and quite long about 23 m 

(75 ft). As an additional safeguard, a barrier should be constructed to prevent access to 

the pond, preferably some type of barrier vegetation. This apartment complex is a new 

development with the majority of the construction occurring in 1997. Therefore, some of 

the landscaping is not complete as yet. There are some shrubs and small trees planted 

around the perimeter, however these do not completely surround the pond. In fact, the 

western slope, which is the longest and steepest approach to the pond, is unfortunately 

the area that does not have barrier vegetation. Since the stormwater runoff is not 

129
 
planned for water contact activities because of high levels of pollutants, e.g. bacteria, 

some proactive effort should be made to discourage the public from making contact with 

the water. 

Peak Reduction Factors (PRF). The dry pond barely reduces the peak outflow rates 

from the contributing area. The 100-year storm’s runoff rate is expected to be reduced 

from 4.3 m3/s (153 cfs) to about 4.1 m3/s (145 cfs) by constriction at the pond’s outlet 

structures and by storage in the pond. 

PRF = 1–Qo/Qi. (10) 

Where: Qo = effluent 

Qi = influent 

The PRF for this event is only 0.05, corresponding to a 5% reduction of the inflow 

hydrograph in the pond. For the 50-year and 25-year storms, the PRFs are 0.06 and 

0.07, respectively. This pond is likely inadequate for any measurable peak flow 

reduction. The pond also likely exceeds the maximum stage of 233.4 m (733 ft). This is 

especially a problem because the water coming through the spillway flows directly onto 

a busy street with high speed traffic. 

Upflow and Critical Settling Velocities. The most important rain events to control for 

water quality improvement are the small events that occur with great frequency, or 

those that produce less than approximately 3.18 cm (1.25 in.) of runoff. For the smaller 

rain events, the DETPOND simulations demonstrated that the retrofitted pond would 

satisfy the maximum up flow velocity (or critical settling velocity) maximum of 0.0004 

mm/s (0.00013 fps) which is necessary for 5 micron particle control. Analyses also 

indicated that removal of 90% of SS was easily obtained on all but the largest rain 

events. The analyses using DETPOND and the 1976 Birmingham rain file (a typical 

year) showed that the annual average suspended control was 86%. 

Pond’s Water Quality Storage. A reasonable goal is to have the pond’s water quality 

storage be equal to the runoff associated with 3.18 cm (1.25 in.) of rain for the land use, 

development, and cover of the watershed served by the pond (Pitt 1998). The 

composite CN for the watershed was estimated to be 75. This corresponds to 

approximately 1 cm (0.40 in) of runoff, from a storm of 3.18 cm (1.25 in.) (Pitt 1987). 

Therefore the minimum active pond storage (between the invert elevation of the lowest 

outlet and the secondary outlet discharge devices) required should be a least 1727 m3 

(1.4 ac-ft). Since the retrofitted pond’s water quality storage is about 826 m3 (0.67 acft), this requirement is not met. 

Pond’s Surface Area Requirements. A pond’s surface area should be sized as a percent 

of watershed’s area based on land use, development, cover, and the particle size 

control desired. The apartment site has residential and woodland land uses. Pitt (1998) 

recommends 0.8% of the watershed area for 5 micron particle control in a commercial 

area (based on NURP, National Urban Runoff Program, studies). Therefore the 

130
 
minimum surface area of the pond should be at least 0.13 ha (0.33 ac.), which is close 

to the normal water surface area of the retrofitted pond of .125 ha (0.31 ac.). 

Possible Problems. Experience and observations of similar ponds in this region of the 

country have shown that the shallow areas near the shore of the retrofitted pond might 

become a breeding area for mosquitoes. The shallow shoreline might allow too much 

aquatic vegetation to grow into the pond’s interior area causing congestion and 

periodically dying and subsequently increasing the BOD. In addition, the aquatic plants 

located on the shallow shelf could grow in and around the outlet and reduce its output 

capacity. Reptiles (including snakes), waterfowl, rodents (i.e., muskrats) might become 

too plentiful in the pond area and become nuisances. Decaying vegetation may result in 

strong odors in the summer, and these odors might have a negative impact on 

residents. 

Background Information Related to Site Evaluation 

The peak inflow hydrograph values were determined by HydroCAD’s soil conservation 

service (SCS) TR-20 methodology. For the site, a SCS Type III rainfall intensity 

distribution frequency (IDF) curve was selected. Rainfall depths for the 100-year, 50­

year, and 25-year storms were approximately 22 cm (8.6 in.), 20 cm (7.8 in.), and 18 cm 

(7.1 in.) respectively. Time of concentration for the watershed was also calculated 

using HydroCAD’s built-in TR-20 methods; Tc = 24.3 min for the apartment complex 

area, and Tc = 33.8 for the Woodland area. 

Land Use, Development, Cover, Soils Type, and CNs. SCS soil maps for the area were 

examined, and it was determined that the site consisted of Nauvoo-Sunlight complex, 

with 15—25% slopes, and Townley silt loam, with 12—18% slopes. The SCS 

Hydrologic Soil Groups for these soils are type B and type C respectively. Research 

conducted at the University of Alabama Birmingham has shown that development, due 

to construction disturbances, compaction, and soil mixing, can significantly reduce the 

actual infiltration rates from those assumed. Therefore, the CN assigned to the 

developed area was for the worst case, type D soil. However, the undeveloped 

woodland area, mostly Nauvoo soil, was assigned a CN based upon the type B soil. 

Therefore for the developed area of 10.5 ha (26.4 ac.), a composite CN of 87 was 

assigned. This composite CN of 87 is a weighted average at the sub-basin CNs. The 

basins include 6.5 ha (16 ac.) of residential land use with one-eighth-acre lots, SCS soil 

type D, and 65% impervious cover (CN = 92), plus the remaining 4.2 ha (10.4 ac.) of 

open lawns with good grass cover (> 75%), and type D soil (CN = 80). For the 

woodland area of 6.1 ha (15 ac.), a CN of 55 was assigned, corresponding to woods 

with good hydrologic condition and type B soils. 

Analysis of Design Storms 

HydroCADTM. The HydroCAD Stormwater Modeling System (version 4.53) was used to 

analyze the pond for hydraulic characteristics. This computer software allows for 

calculating hydrographs, based upon design storm and watershed characteristics, and 

then routing these through a drainage system composed of subcatchments, reaches, 

131
 
and ponds. The program does not consider the dead storage below the first outlet, 

assuming that this is always full of water, therefore the flow characteristics of both the 

existing pond and the retrofitted pond are identical. 

The subcatchment component is used to model a given drainage area or watershed. In 

this case, there are two subcatchments: subcatchment 1 refers to the 10.5 ha (26.4 ac.) 

of the apartment complex, and subcatchment 2 consists of the 6.1 ha (15 ac.) of 

woodland area that drains to the complex. The program uses built-in SCS TR-20 

hydrology methods for determining the hydrograph characteristics. Next, the 

hydrograph is routed through a series of defined reaches and/or ponds. In this case, 

there are two hydrographs, one from each of the two subcatchments, which are routed 

through a single pond. 

The pond component of this model is described using a stage v. surface area curve, 

shown in Figure 5.3.c-1 and Table 5.3.c-1. In addition, the model requires descriptions 

of the outlet structures. 

Elevation v. Surface Area 

Stonecrest Dry Detention Pond 

0.32 

0.28 

0.24 

0.20 

0.16 

0.12 

0.08 

0.04 

720 721 722 723 724 725 726 727 728 729 730 731 732 733 

Water Surface Elevation (ft) 

Figure 5.3.c-1. Stage v. Surface Area Curve. Surface Area (ac) 

132
 
Table 5.3.c-1. Outlet Device Descriptions 

# Route Invert Outlet Device 

1 Primary 219.5 m (720 ft) 61 cm (24 in) culvert 

n=0.013, length=56.4 m (185 ft), 

slope=0.02 m/m, Ke=0.5, Cc=0.9 

2 to #1 219.5 m (720 ft) 30.5 cm (12 in) orifice 

3 to #1 221.6 m (727 ft)	 sharp-crested rectangular weir 

length = 3.7 m (12 ft) 

height = 20.3 cm (8 in) 

(square concrete box with cap) 

4 Secondary 223.1 m (732 ft)	 3 m (10 ft) broad-crested 

rectangular weir emergency 

spillway 

Ke = 0.5 represents the entrance loss for a flush entrance
 

Cc = 0.9 represents the coefficient of discharge for the culvert
 

The HydroCAD simulations were run for three, 24 h, SCS Type III design storm 

frequencies: 25-year 18 cm (7.1 in.), 50-year 20 cm (7.8 in.), and 100-year 22 cm (8.6 

in.). Table 5.3.c-2 summarizes the model’s output for these three storms, showing the 

hydrographs peaks and volumes. As expected, an examination of the last two columns 

shows that the most significant contribution to the hydrograph flowing into the pond 

comes from the apartment complex area. The surface area of the complex is larger 

than that of the woodlands and land use, soil characteristics, and other hydrologic 

characteristics of the complex all contribute to generating much more and rapid runoff 

per unit area. 

Table 5.3.c-2. Subcatchment Summaries for Design Storms 

Subcatch- Description Design Storm Rainfall Peak Flow Volume 

ment No. Frequency (yr) (cm) (in) (m3

/s) (cfs) (m3

) (ac-ft) 

1 apartment 25 18 7.1 2.91 102.6 13,310 10.79 

complex 

2 woodland 25 18 7.1 0.54 19.13 3,035 2.46 

1 apartment 50 20 7.8 3.24 114.4 14,850 12.04 

complex 

2 woodland 50 20 7.8 0.67 23.59 3,690 2.99 

1 apartment 100 22 8.6 3.62 127.9 16,615 13.47 

complex 

2 woodland 100 22 8.6 0.82 28.94 4,490 3.64 

Next the hydrographs were routed through the pond. Table 5.3.c-3 summarizes the 

results. As previously mentioned, the peak reductions are quite low (5—7%), and the 

peak discharge lag is only a few minutes. The predicted peak water surface elevation in 

the pond is higher than the maximum elevation in the pond, 223 m (733 ft) (column 3), 

133
 
indicating flooding. This is a dangerous situation because it means that the water is 

flowing uncontrollably over the dam. This could damage the emergency spill way and 

cause erosion of the dam itself. These events occur even at the lowest storm frequency 

modeled 25-year. It would appear that the pond is inadequate for the amount of runoff 

generated by these storms. 

Table 5.3.c-3. Pond Results of HydroCAD simulations 

Design Peak Elev.** Peak Storage Peak Q in Peak Q out Peak Q outlet† Peak Q emerg ‡ Atten. Lag 

Event* Time 

(m) (ft) m 3 ac-ft. m 3/s cfs m 3/s cfs m 3/s cfs m 3/s cfs % (min) 

25-yr. 233.5 733.4 2,603 2.11 36.1 118.3 33.6 110.2 15.4 50.5 18.2 59.7 7 8.1 

50-yr. 233.7 733.8 2,726 2.21 40.9 134.2 38.4 126.1 15.6 51 22.9 75.2 6 7.2 

100-yr. 223.8 734.1 2,850 2.31 46.5 152.6 44.2 144.9 15.7 51.6 28.5 93.4 5 5.8 

* Design storms are type III 24 h for Shelby County (SCS methods). 

** Flood elevation is at 223.4 m (733 ft). 

† Peak flow through the first and second outlets to a 0.6 m (2 ft) diameter culvert. 

‡ Peak flow in the emergency spillway (flowing onto Bowling Drive). 

DETPOND Modeling. DETPOND generally uses a simplified triangular hydrograph for 

individual storms in long-term simulations (several decades) for water quality 

evaluations, so this model is not suitable for modeling synthetic rainfall distributions, 

such as those appropriate for design storms. However, HydroCAD is capable of these 

types of simulations, and one can input the hydrograph generated by HydroCAD into 

DETPOND using the “user defined hydrograph” option. Therefore, the 25-year, 50-year, 

and 100-year hydrographs generated using HydroCAD’s TR-20 methods were input, 

and these results are presented in the Table 5.3.c-4. A comparison of the hydraulic 

results from HydroCAD (Table 5.3.c-4) shows that these results are similar. For 

instance during the 25-year event, HydroCAD calculated a maximum flow rate into the 

pond of 36.1 m3

/s (118.3 cfs), while DETPOND’s calculations resulted in 36.0 m3/s 

(118.0 cfs). Even under these worst-case conditions, the pond is removing 

approximately 50% of the SS. 

134
 
Table 5.3.c-4. DETPOND Summary for Design Storms 

Design Max. Stage Max. Inflow Max. Outflow Max particle Avg. Min Particle % SS 

Event (m) (ft) (m3

/s) (cfs) (m3/s) (cfs) size discharged 

(μm) 

Size Controlled 

(μm) 

Removed 

25 yr 223.6 733.5 36.0 118.0 30.5 100.1 95 26.5 50.8 

50 yr 223.7 734.0 40.9 134.2 36.0 118.0 95 28.1 48.6 

100 yr 223.9 734.5 46.5 152.6 41.6 136.6 95 29.7 46.7 

Analysis Using True Rainfall Characteristics 

DETPOND evaluates long-term continuous simulations using many years of actual 

rainfall data. Rain files are created by determining the rain depth, rain duration, and 

interevent duration for multiple storms. DETPOND then routes a simple, triangular 

hydrograph for each event, and routes it through the pond to evaluate the SS removal. 

DETPOND simulations were conducted using rain files created from the 1976 

monitoring data, and also the 1952 through 1989 rain record. The 1976 file is 

representative of a typical year, and was used to enable more model runs because of 

shorter run time. The larger file was then run to continuously evaluate the pond based 

on almost forty years of actual storm data. There are only 23 events, out of a total of 

4,107 in the large 40 year rain file in which the pond stage rises to the level of the 

second outlet. All of the DETPOND simulations were conducted using the retrofitted 

pond, as water quality benefits in the existing dry pond were assumed to be negligible. 

The results of the simulations using the 1976 file are presented in Table 5.3.c-5. 

Table 5.3.c-5. Water Quality Output Summary for 1976 Rain File 

Inter- Max Pond Flow­ Approx.

Stage weighted Part. Rain Depth Rain Event Rain Intensity Peak Event 

Statistic (cm) (in) Duration Duratio 

(cm/h) (m) (ft) Particle Res. Reduction Flushing 

(h) n (ds) (in/h) 

Size Control* Factor Ratio** 

Mean 1.27 0.50 12.01 1.81 0.10 0.04 1.94 6.30 4.26 86.3% 0.07 1.75 

Std. Dv. 1.91 0.75 10.77 2.36 0.15 0.06 0.16 0.51 4.23 13.5% 0.07 2.41 

COV 3.84 1.51 0.90 1.30 3.76 1.48 0.03 0.08 0.99 0.2% 1.00 1.37 

Minimu 0.03 0.01 1.00 0.00 0.00 0.00 1.84 6.00 0.00 57.3% 0.01 0.00 

m 

Max. 9.75 3.84 45 11.68 0.79 0.31 2.73 8.84 15.70 100% 0.31 9.34 

* Approximate Particle Residue Control (SS). 

** 	 Event flushing ratio – the storm runoff volume divided by the water volume contained in the pond at the beginning 

of the event. 

The control for the 5 �m particles corresponds to approximately 90% SS control based 

on the mean values of the columns of Flow Weighted Particle Size and Approximate 

135
 
Particle Control. On average, in a typical year, the pond will collect particle sizes 4.26 

microns and greater in size. The average rain depth is 1.27 cm (0.5 in.), and the 

average duration is 12 h. For the smallest storms, the pond is achieving greater than 

95% control, and for the largest storm in 1976, the SS control is still 57%. 

Figure 5.3.c-2 shows the maximum pond stage. Axis labels denote the elevation above 

the pond bottom (1.8 m [6 ft] corresponds to 219.5 m [720 ft] msl elevation, the invert of 

the first outlet device) versus the percent particle control. There is an expected trend; 

the control decreases with increasing stage. 

Redesigned Pond, Bham76 

Particle Control (%) 

100 

95 

90 

85 

80 

75 

70 

65 

60 

55 

6 6.5 7 7.5 8 8.5 9 

Maximum Pond Stage (ft above bottom) 

Figure 5.3.c-2. Pond Stage v. Particle Residue Control 

Figure 5.3.c-3 shows the water quality performance of the retrofitted pond (percent 

particulate control) versus the rain depth in inches. There is considerably more scatter 

in this plot, but one can still observe a trend. Generally, control decreases as rain depth 

increases, as expected. The results are similar for Figure 5.3.c-4 depicting control 

versus rain intensity. 

136
 
Redesigned Pond, Bham76 

Particle Control (%) 

100 

95 

90 

85 

80 

75 

70 

65 

60 

55 

0 0.5 1 1.5 2 2.5 3 3.5 

Rain Depth (in) 

Figure 5.3.c-3. Rain Depth v. Particle Residue Control 

Redesigned Pond, Bham76 

100 

95 

90 

Particle Control (%)

85 

80 

75 

70 

65 

60 

55 

0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 

Rain Intensity (in/hr) 

Figure 5.3.c-4. Rain Intensity v. Particle Residue Control 

137
 

4 
Long-term Simulation of 1952-1989 Rains 

Table 5.3.c-6 shows the statistics for the 4,107 rain events that occurred in Birmingham 

from 1952-1989. Notice that the minimum and maximum values are different than those 

from the 1976 simulations, but the mean values are quite similar. The most important 

value in this table is the mean particle control, 80%. For 1976, the control was 

approximately 86%. This is not significantly different. With close to forty years of rains, 

the redesigned pond still averages 80% SS removal. 

Table 5.3.c-6. Water Quality Output Summary for 1952—1989 Rain File 

Inter- Max Pond Flow­ Approx. 

Stage weighted Part. Res. Rain Depth Rain Event Rain Intensity Peak Event

Particle Control* Reductio Flushing Statistic (cm) (in) Duration Duration 

(cm/h) (in/h) (m) (ft) Size (%) (h) (ds) n Factor Ratio** 

Mean 1.27 0.50 6.31 2.57 0.10 0.04 1.94 6.30 6.43 80.13 0.13 1.91 

Std. Dv. 1.91 0.75 6.88 3.54 0.15 0.06 0.16 0.51 4.91 14.36 0.10 2.44 

COV 3.84 1.50 1.09 1.38 3.76 1.48 0.03 0.08 0.76 0.18 0.76 1.28 

Minimu 0.03 0.01 1.00 0.00 0.00 0.00 1.84 6.00 0.20 48.50 0.00 0.00 

m 

Max. 34.4 13.5 93 44.31 0.79 0.31 2.73 8.84 23.2 99.8 0.57 9.96 

9 8 

* Approximate Particle Residue Control (SS). 

** Event flushing ratio – the storm runoff volume divided by the water volume contained in the pond at the beginning of the 

event. 

Conclusions 

As described above, retrofitting a dry pond to a wet pond can provide improvements to 

the overall benefits of the device. Average SS removal for particles of 5 mm or greater 

is expected to approach 80% removal annually. Additionally, a combination of water 

quality and flow analysis computer programs can be used together to evaluate and 

design effective stormwater controls and SW control structures can provide both water 

quality protection and flood control. 

138
 
5.4 Retrofitting WWF Storage Tanks for DWF Augmentation 

The operation of a WWTF requires the accommodation of wide fluctuations in flow due 

to naturally occurring variations in the generation of wastewater and the associated 

effects of I/I. Diurnal variations in flow rates are typically observed with higher flow rates 

occurring during the daytime when water use is high and lower flow rates at night when 

water use is low. 

Wastewater equalization can dampen these diurnal variations to achieve a relatively 

constant hydraulic loading to the WWTF. Equalization basins are normally located 

within the WWTF following grit and screening facilities; however, equalization is 

sometimes located in the collection system to relieve overloaded trunk sewers. 

Equalization can relieve overloaded treatment facilities in existing treatment plants and 

can improve the efficiency, reliability, and operability of these facilities. Equalization 

may also be a more cost-effective solution to modifying existing facilities. Retrofitting 

existing WWF storage tanks to serve a second function as DWF equalization tanks may 

prove to be a cost-effective alternative to the modifying existing treatment facilities or 

the construction of new equalization tanks at the treatment plant. Two desktop 

analyses will be evaluated to demonstrate this: 

• Retrofit of the existing Erie Boulevard Storage System (EBSS) in Syracuse, NY 

•	 Retrofit of the existing Spring Creek Auxiliary Wastewater Pollution Control Plant in 

New York, New York. 

5.4.a. Erie Boulevard Storage System, Syracuse, New York 

Facility Description 

The EBSS was constructed as part of the Best Management Practice (BMP) 

improvements in 1982. The existing large diameter storm sewer running underneath 

Erie Boulevard 2x3 m (7.5x10.5 ft) was retrofitted with automated sluice gates and level 

sensors to entrap CSOs that discharge to this structure. These discharges are 

temporarily stored in the EBSS until the Syracuse Metropolitan Wastewater Treatment 

Facility (METRO) has capacity to accept the WWF. The 18,950 m3

 (5 MG) volume of 

the EBSS can adequately contain the discharge from the 90 percentile storm in its 

tributary area. Operational problems with control equipment prevented the system from 

functioning as intended. Reactivation of the EBSS as a WWF storage facility is targeted 

for completion by July 2002. 

As shown in Figure 5.4.a-1, the EBSS receives flow from three major sources: 

1.	 Burnet Avenue Trunk Sewer (BATS) overflows, including the James Street Relief 

Sewer (JSRS). 

2. Fayette Street Trunk Sewer (FSTS) overflows. 

3.	 Storm Sewers at the upstream end of the system as part of the natural drainage 

basin. 

139
 
Figure 5.4.a-1. EBSS Layout 

140
 
The EBSS will be automatically controlled through the use of state-of-the-art SCADA 

control system. This system consists of aboveground sluice gate control structures for 

each of the four control gates that operate based on level sensors in the influent 

sewers. The SCADA system also controls operation of the inflatable dam that regulates 

flow from the BATS, the major source of flow to the EBSS. 

Three sluice gates (Gates 1, 3, and 4) on the EBSS establish maximum water surface 

elevations as shown on Figure 5.4.a-1. Gate Chamber 2 is located on the regulator 

sewer that connects the EBSS to the main interceptor sewer (MIS) that conveys flow to 

METRO. This control gate operates based on the water level in the MIS. 

Upon the initiation of WWF into the EBSS, it is filled sequentially through the closure of 

Gate Chambers 1, 3, and 4. Flow fills each basin sequentially until the 18,950 m3 (5 

MG) volume is exceeded, at which time, flow exits the EBSS through the high-water 

overflow at Gate Chamber 1. Gate Chamber 2 is closed except for dewatering of the 

EBSS following a wet-weather event as described below. 

Stored volume behind Gate Chamber 1 is dewatered first by opening Gate Chamber 2 

to convey flow to the MIS. Gate Chamber 3 is dewatered following dewatering of Gate 

Chamber 1 by opening Gate Chamber 3, and then Gate Chamber 4 is opened allowing 

the stored volume to flow into the lower chambers to the MIS for ultimate treatment at 

METRO. Gate Chamber 1 remains closed during this phase. Gate Chamber 1 is 

opened upon final dewatering of the system and EBSS is returned to standby mode. 

This facility in its intended wet-weather operational mode will be used to evaluate the 

hypothetical potential for retrofitting the EBSS to serve a second function for DWF 

equalization. Equalization of DWF in the EBSS could prove to be beneficial to the 

treatment plant by eliminating the diurnal flow variability, providing METRO a relatively 

constant DWF component from the EBSS tributary area. 

Proposed Retrofit 

A portion of the DWFwill be directed from the BATS to the EBSS through the JSRS by 

the use of regulator structure located in the BATS immediately downstream of the 

JSRS. This structure is proposed as part of the EBSS reactivation to direct WWF into 

the EBSS. This structure contains a bottom-outlet orifice device to allow the DWF to be 

conveyed downstream through the BATS. The proposed retrofit will modify this orifice 

to allow only a portion of the dry-weather to pass downstream while regulating the 

higher DWF into the EBSS. DWF in excess of the orifice capacity will be directed into 

the EBSS where it will be stored and regulated to the MIS using a new vortex valve that 

would be retrofitted into Gate Chamber 2, for ultimate treatment at METRO. 

The storage volume between Gate Chambers 1 and 3 targeted for use for DWF 

equalization is 8,680 m3

 (1.5 MG) based on the storage volume below the invert 

elevation of the JSRS. This will allow the remaining 13,250 m3 (3.5 MG) of the 18,950 

m (5 MG) EBSS capacity (2,650 m3 (0.7 MG) remaining capacity between Gate 

141
 

3
Chambers 1 and 3 and 10,600 m3

 (2.8 MG) upstream of Gate Chamber 3) to be 

available for storage of WWF. 

Monitoring of the BATS downstream of the JSRS conducted from July 29, 1998 through 

August 6, 1998 found the DWF to vary between 0.16-0.69 m3/s (5.7-12.8 cfs) with an 

average of 0.22 m3/s (7.6 cfs). It is proposed to divert DWF in excess of 0.15 m3/s (5.2 

cfs) to the EBSS for equalization to provide, on average, 8,680 m3

 (1.5 MG) of flow to 

be equalized per day utilizing the capacity of the previously described DWF volume. 

Stored DWF will be routed to METRO through a new vortex valve that will be retrofitted 

into the existing Gate Chamber 2 with a capacity of 0.065 m3

/s (2.3 cfs). This low 

release rate will allow effective equalization of DWF and utilization of the proposed 

8,680 m3

 (1.5 MG) dry-weather storage volume. Gate Chamber 2 is directly connected 

to the MIS and will serve as the drain of the entire EBSS during wet-weather operation. 

Stormwater Management Model Results 

The EPA SWMM Version 4 assessed the impact of the retrofit by comparing the use of 

the EBSS under pre- and post-retrofit conditions. A simplified model was simulated 

hourly rainfall, based upon a previously calibrated model. 

The data consisted of 30 years of hourly rainfall data form the NWS station at Hancock 

Field in Syracuse, NY for the years from 1962-1991 and were used as input to the 

simplified RUNOFF model. The projected flows were used to simulate long-term 

overflow using the TRANSPORT block of SWMM. The TRANSPORT model was 

constructed as a simplified network consisting of overflows and regulator pipes. The 

simplified model was used to project the annual wet-weather capture from the EBSS 

tributary area. 

Table 5.4.a-1 demonstrates that the EBSS operated in a wet-weather mode provides a 

considerable annual capture of WWF. The retrofitting of the EBSS for DWF 

equalization reduces the ability to capture and retain WWF. Table 5.4.a-1 shows that 

this retrofit reduces the overall WWF capture from 80% to 75%. 

142
 
Table 5.4.a-1. Average Annual Capture Volume from the EBSS Under Pre- and PostRetrofit Conditions (1962 – 1991) 

Average Annual Total 

WWF in the Combined 

Sewer System 

Average Annual 

WWF Captured Percent 

Volume 

EBSS CONDITION (m3

) (MG) (m3

) (MG) Capture 

Existing Wet-Weather 1,071,465 283 859,450 227 80% 

Operation 

Dry-Weather Retrofit 1,071,465 283 802,650 212 75% 

Operation(1) 

(1) Assumes dry-weather portion of storage is fully utilized during wet-weather events to 

represent worst case 

Design Considerations 

Retrofitting the EBSS to provide equalization of a portion of the DWF from the BATS 

required careful consideration of several key design factors, as follows: 

• Hydraulic Loadings 

• Flushing Requirements 

• Operational and Maintenance Requirements 

These design factors are described in further detail in this section. 

Hydraulic Loadings. It is proposed to divert DWF in excess of 0.15 m3/s (5.2 cfs) to the 

EBSS for equalization. This would result in an average of 8,680 m3 (1.5 MG) of flow to 

be equalized per day. DWF in excess of the available 8,680 m3

 (1.5 MG) storage 

volume will be directed to METRO through the existing BATS. 

Operational Requirements. The only adjustable control point necessary for this retrofit 

is a pinch valve located on the dry-weather orifice device. The dry-weather orifice will 

be automatically adjusted when flow in the BATS exceeds the maximum DWF of 0.69 

m3

/s (12.8 cfs) or when the dry-weather storage volume is fully utilized. This will be 

done by fully opening the pinch valve on the dry-weather orifice to allow the entire DWF 

to be conveyed downstream through the BATS. This will be controlled through the 

EBSS SCADA system. The pinch valve will be engaged in the normal operational mode 

to only allow passage of the lower portion of the DWF. 

Flushing Requirements. Accumulated solids deposition will increase during equalization 

of DWF, however, the layout of the EBSS serves to provide considerable flushing 

following wet-weather events. This is accomplished by staged release of Gate 

143
 
Chambers 3 and 4 to provide a wave of water to direct accumulated solids downstream 

for conveyance to METRO through the MIS. This system will be an effective method for 

the removal of accumulated solids from the DWF equalization. 

Operation and Maintenance Requirements. The use of EBSS for DWF equalization will 

result in increased O&M costs. The following list summarizes the areas of expected 

O&M cost increases: 

• Labor to perform inspection of the facilities 

• Power for the control system 

Table 5.4.a-2 summarizes the expected increase in annual O&M costs resulting from 

the use of the EBSS for DWF equalization. 

Table 5.4.a-2. Estimated O&M Costs 

Estimated Annual 

Description Cost 

Process Control Adjustments $3,000 

Clean-Up of Wet-Weather Treatment $10,000
 

Structures 

Power $2,000 

Total Estimated O&M Cost $15,000 

Project Costs 

The projected capital costs for retrofitting the EBSS for DWF equalization are provided 

in Table 5.4.a-3. 

Table 5.4.a-3. Estimated EBSS Retrofit Capital Costs 

Description Estimated Cost(1) 

Flow Diversion Structure Retrofit $10,000 

Flow Regulator Structure Retrofit $15,000 

Electrical Controls $35,000 

Instrumentation and SCADA System $55,000 

Total Project Cost $140,000 

(1) Includes Design Engineering, Construction Engineering, Legal, and Fiscal 

Expenses, does not include contingencies and right-of-way costs 

144
 
5.4.b. Spring Creek AWCP, New York, New York 

The Spring Creek Auxiliary Water Pollution Control Plant (AWPCP) is a CSO wetweather detention facility with approximately 49,200 m3 (13 MG): 37,900 m3

 (10 MG) in 

basin storage and 11,400 m3 (3 MG) in influent barrel storage. The details of the Spring 

Creek facility are provided in Section 5.2 of this document. 

Proposed Retrofit 

A portion of the DWF will be directed from the Autumn Avenue regulator in Brooklyn to 

basin 6 to minimize the diurnal fluctuations of DWF from Autumn Ave. regulator 

drainage area that is tributary to the 26 Ward WPCP. The DWF will be conveyed to 

AWPCP to allow a specified level of DWF to be conveyed to the 26 Ward WPCP. 

The storage volume in basin 6 targeted for use for DWF equalization is 3,790 m3 (1.0 

MG) or approximately one-half of the storage volume. This will allow the remaining 

34,000 m3

 (9.0 MG) of the 37,900 m3 (10 MG) AWPCP capacity to be available for 

storage of WWF. 

O’Brien & Gere Engineers previously performed an extensive evaluation of the 26 Ward 

WPCP throughout the past 15 years. Their work has shown the average DWF from the 

Autumn Avenue regulator to vary between 1.06-2.07 m3/s (37-73 cfs) with an average of 

1.5 m3

/s (53 cfs). It is proposed to divert DWF in excess of 1.33 m3/s (47 cfs) to 

AWPCP Basin No. 6 for equalization to provide, on average, 3,790 m3 (1.0 MG) of flow 

to be equalized per day. Stored DWF will be routed to the 26 Ward WPCP through a 

new dry-weather pump station with a capacity of 0.28 m3

/s (10 cfs). This low release 

rate will allow effective equalization of DWF and utilization of the proposed 3,790 m3 

(1.0 MG) dry-weather storage volume. 

The proposed retrofit will consist of a physical modification to the existing diversion 

structure by installation of sluice gate on the 1.5 m (60 in.) diameter dry-weather outlet 

of the regulator structure. This sluice gate will be controlled by a flow meter located 

downstream of the sluice gate in the 1.5 m (60 in.) diameter interceptor sewer to limit 

flow to 1.33 m3

/s (47 cfs). The regulator chamber will also be modified to include a weir 

structure to convey DWF in excess of 47 cfs to Bay Number 6 for equalization. Level 

sensors in the dry-weather equalization tank will also be used to control the dry-weather 

sluice gate when the 1 MG volume is utilized. 

The dry-weather sluice gate will also be designed to open fully when sewage overflows 

the Autumn Ave. regulator structure into Spring Creek to maximize flow to the treatment 

plant by the used of a level sensor located at the invert elevation of the overflow weir. 

145
 
Stormwater Management Model (SWMM) Results 

The SWMM Version 4 assessed the impact of the retrofit by comparing the use of the 

Spring Creek AWPCP under pre- and post-retrofit conditions. A simplified model was 

use to simulate hourly rainfall, based upon a previously calibrated model. 

The data consisted of hourly rainfall data from the National Oceanographic and 

Atmospheric Administration rain gauge No. 305803 at J.K.F. International Airport in New 

York, NY for the 1985. This data was used as input to the simplified RUNOFF model. 

The projected flows were used to simulate long-term continuous simulation using the 

TRANSPORT block of SWMM. The TRANSPORT model was constructed as a 

simplified network consisting of the overflow and regulator pipes. The simplified model 

was used to project the annual wet-weather capture from the Spring Creek AWPCP 

under pre- and post-retrofit conditions. 

Table 5.4.b-1 demonstrates that the AWPCP presently discharges a considerable 

amount of WWF to Spring Creek. The retrofitting of the AWPCP for DWF equalization 

reduces the ability to capture and retain WWF only to a small extent. Table 5.4.b-1 

shows that this retrofit increases the annual WWF by 12% or approximately 50,000 m3 

per year (13.2 MG per year). 

Table 5.4.b-1. Wet-Weather Discharge Volume from the Spring Creek AWPCP Under 

Pre- and Post-Retrofit Conditions for 1985 

Wet-Weather 

Discharge for 1985 

Spring Creek Storage Volume (m3) (MG) 

Existing (37,850 m3

 [10 MG]) 433,005 114.4 

Retrofitted (34,065 m3 [9 MG]) (1) 482,967 127.6 

Increase In Wet-Weather Discharge 49,962 13.2 

Percent Increase 12% 12% 

(1) Assumes dry-weather portion of storage is fully utilized during wet-weather 

events 

Design Considerations 

Retrofitting the AWPCP to provide equalization of a portion of the DWF from the 

Autumn Avenue regulator required several design factors: 

• Hydraulic Loadings 

• Operational Requirements 

• Flushing Requirements 

• Operational and Maintenance Requirements 

146
 
These design factors are described in further detail in this section. 

Hydraulic Loadings – It is proposed to divert DWF in excess of 1.33 m3

/s (47 cfs) to 

AWPCP for equalization. This would result in an average of 8,680 m3 (1 MG) of flow to 

be equalized per day. DWF in excess of the available 8,680 m3

 (1 MG) storage volume 

will be directed to the 26 Ward WPCP. 

Operational Requirements – The two primary control points for this retrofit are the new 

DWF sluice gate that will be installed in the Autumn Ave. regulator and the new DWF 

pump station that will be used to convey equalized flow to the 26 Ward WPCP. 

The dry-weather sluice gate will be automatically adjusted when flow in the 1.5 m (60 

in.) diameter interceptor exceeds 1.33 m3/s (47 cfs) or when the 1 MG dry-weather 

storage volume is fully utilized. 

New level sensors will be used to control operation of the DWF pump. These controls 

will be set to energize the pump station when flow is sensed in the storage tank and to 

turn the pump off when the tank is empty. 

Flushing Requirements – Accumulated solids deposition will increase during 

equalization of DWF, however, a flushing system will remove accumulated solids 

following wet-weather events. The design of new flushing system for the Spring Creek 

AWPCP is presently being evaluated to provide more efficient removal of accumulated 

solids. The selected system will be an effective method for the removal of accumulated 

solids from the DWF equalization. 

Operation and Maintenance Requirements – An increased O&M burden will be created 

for the use of Spring Creek for DWF equalization. The following list summarizes the 

areas of expected O&M cost increases: 

• Electrical Power to operate the dry-weather pump station and the orifice control 

• Labor to perform inspection of the facilities 

• Labor to perform flushing of accumulated solids 

Table 5.4.b-2 summarizes the expected increase in annual O&M costs resulting from 

the use of Spring Creek for DWF equalization. 

Table 5.4.b-2. Estimated O&M Costs 

Description Estimated Annual Cost 

Electrical Power $3,000 

Labor (Maintenance of Facilities) $10,000 

Labor (Tank Flushing) $25,000 

Total Estimated O&M Cost $38,000 

147
 
Project Costs 

The projected capital costs for retrofitting Spring Creek for DWF equalization are 

provided in Table 5.4.b-3. 

Table 5.4.b-3. Estimated Spring Creek DWF Retrofit Capital Costs 

Description Estimated Cost1 

Regulator Structure Retrofit $205,000 

DWF Pump Station and Force Main $170,000 

Process Control System $45,000 

Instrumentation and SCADA System $65,000 

Total Project Cost $485,000 

1 Includes Design Engineering, Construction Engineering, Legal, and 

Fiscal Expenses, Does not include contingencies and right-of-way costs 

Conclusions 

Retrofitting the EBSS and Spring Creek AWPCP were evaluated for the purpose of 

serving a dual role as wet-weather and dry-weather equalization facilities without 

drastically reducing the ability of the storage tank to capture WWF. The primary benefit 

of equalization provided by the DWF storage is the dampening of peak flow rates that 

would otherwise result at the treatment plant. This retrofit will increase the treatment 

capacity for a treatment plant that is operating at or near design capacity and will also 

allow the primary and secondary treatment facilities to operate at a more uniform rate to 

provide consistent treatment efficiency throughout a typical day. However, there is 

adequate hydraulic and biological capacity the Syracuse METRO and 26 Ward 

treatment facilities. 

This retrofit option may also provide a reduction in operation of high flow pumps, where 

appropriate, as a result of eliminating the high diurnal flows. This could result in a 

reduction in O&M at the treatment plant; however, the O&M required at the equalization 

facility may offset any benefit at the treatment plant. 

148
 
5.5 SSO Control Using Storage 

5.5.a. Hypothetical Overflow Retention Facility Retrofit to an Existing Sanitary 

Sewer System 

This desktop analysis involves a hypothetical retrofit to a sanitary sewer system to 

eliminate SSOs from an existing regulator structure. This desktop analysis was 

developed from a real life project from a confidential source. Overflow from this 

regulator occurs during and following rainfall events due to a well-established I/I 

problem. The I/I problem was assessed as part of a sewer system evaluation survey 

(SSES) that identified and quantified various sources of I/I with the collection system. A 

rehabilitation program was implemented in accordance with the findings of the SSES to 

provide removal of the primary I/I sources. This program also requires the elimination of 

the existing regulator overflow. 

Sewer system modeling was performed using the Environmental Protection Agency’s 

SWMM Version 4. The previously calibrated model was expanded to identify flow 

conditions at the regulator following the implementation of the I/I rehabilitation program. 

The model was used to identify peak overflow rates, volumes and hydraulic grade lines 

for various overflow abatement alternatives. 

Project Approach 

A 0.9 m (36 in.) diameter interceptor and an 0.5 m (18 in.) diameter interceptor enter the 

regulator chamber. A 0.9 m (36 in.) diameter changes to a 1 m (42 in.) diameter 

interceptor then exits the chamber and continues to the WWTF. The existing 

interceptor and other downstream piping configurations restrict the flow to the treatment 

plant. An 0.5 m (18 in.) diameter overflow also exits the chamber that discharges to the 

receiving water body when wet-weather events cause the chamber to overflow. 

The regulator overflow was likely developed to relieve a persistent flooding problem in 

the vicinity of the regulator. The interceptor network in the vicinity of the regulator 

presently surcharges during the two-year design storm resulting in approximately 

12,100 m3

 (3.2 MG) of surface flooding. The existing overflow relieves surcharge and 

minimizes flooding. The peak hydraulic grade line cannot increase beyond surcharge 

with the overflow closed. Closing the overflow would result in pronounced basement 

and street flooding in the areas upstream of the existing overflows and would not 

increase the peak flow to the treatment plant from the interceptor. The high head losses 

are associated with the overflow and limitations of the interceptor network that includes 

two double-barrel siphons downstream of the regulator overflow. The alternative of 

constructing a new sewer parallel to the existing sewer to convey flow to the treatment 

plant was not considered feasible given long length of sewer, deep rock excavation, to 

long siphons, and a shoreline construction. 

The two-year frequency, two-hour duration design storm was selected in order to 

determine what control was necessary in terms of treatment, transportation or storage 

of the regulator overflows. Based on the SWMM model, during a two-year storm, a 

149
 
peak rate of approximately 28,000 m3

/d (7.4 MGD) overflows to the receiving water with 

a total overflow volume of 6,400 m3

 (1.7 MG). One recommended abatement 

alternative is to use a 6,400 m3 (1.7 MG) ORF to eliminate overflows up to the two-year 

design storm and to treat overflows in excess of the storage capacity with high-rate 

disinfection. 

The WWTF is presently permitted to treat 64,000 m3

/d (17 MGD). The monthly average 

flow during a two-year monitoring period of the SSES was 76,000 m3

/d (20 MGD). As 

the plant already exceeds its flow limit, recommendations to increase the peak WWF to 

the plant were not made, but to store the WWF during wet events with a slow release to 

the WWTF following the storm. 

Overflow Retention Facilities 

The proposed retrofit consists of a 6,400 m3 (1.7 MG) ORF located between the 

regulator chamber and the existing outfall to capture the flow associated with the twoyear design storm. Overflow events producing less than 6,400 m3

 (1.7 MG) are 

retained and returned to the treatment plant when influent flows subside. Overflow 

events producing in excess of the available storage volume are disinfected using 

chlorine and dechlorinated before being discharged to the receiving water. The retained 

volume of 6,400 m3

 (1.7 MG) is returned to the treatment plant through the interceptor 

sewer utilizing an overflow return pump. 

The overflow retention facilities include the following components: 

ORF Tanks. The ORF tanks will consist of an in-ground reinforced concrete tank with a 

dimension of 30.5 m (100 ft) wide by 73.2 m (240 ft) long with a 3 m (10 ft) depth to 

provide a useable retention volume of 6,400 m3 (1.7 MG). 

Overflow Return. A 10.2 cm (4 in.) overflow return pump will be used to return the 

captured volume to the WWTF through the existing interceptor sewer within 48 h after 

the occurrence of an overflow event. 

Disinfection. The disinfection system will utilize sodium hypochlorite introduced into a 

contact tank sized for 5 min detention time at peak flow rate of 28,000 m3

/d (7.4 MGD) 

and requires the use of mechanical mixers to induce energy to provide effective 

disinfection. The mixing velocity gradient and chlorine dose were calculated on the 

basis of estimated influent and effluent fecal coliform levels for sanitary sewer 

overflows. The high-rate disinfection system was sized by estimating disinfectant 

dosage and mixing energy required to assure the inactivation of fecal coliform. Based 

on the optimum velocity gradient calculated, horsepower requirements for mixers were 

determined. From this analysis, it was determined that an optimum chlorine dosage of 

20 mg/l and approximately 1.1 kW (1.5 HP) of total mixing energy are required to insure 

proper disinfection. A 5 cm (2 in.) disinfection tank pump will be used to return the 

disinfection tank volume to the WWTF through the existing interceptor sewer within 48 h 

after the occurrence of an overflow event. 

150
 
Dechlorination. Dechlorination, which reduces the chlorine residual prior to discharge, 

will be accomplished by injecting sodium bisulfite in a 10 m3 (2,600 gal) tank with a 

high-rate mixer to provide detention time of 30 s at a peak flow of 28,000 m3/d (7.4 

MGD). The dechlorination facilities are metered automatically as a function of overflow 

flow rate to minimize chlorine residual. 

Overflow Metering. Overflows will be metered following discharged from the 

dechlorination tank through an open channel-type metering device located in the outfall 

pipe. 

Control Building. A 63.2 m2 (680 ft2

) structure will be provided as the control building for 

this facility. This structure will include process control equipment and be used to house 

the chemical storage tanks. 

Stormwater Management Model Results 

The EPA SWMM Version 4 was used to assess the benefit of the retrofit by comparing 

annual WWF volume presently discharged to the volume captured by the retention tank. 

A simplified model was developed to simulate hourly rainfall for the hypothetical 

collection system 

The data used for these projections consisted of 15 years of hourly rainfall data for the 

years from 1961-1985. These data were used as input to the simplified RUNOFF model 

that projected flows that were used to simulate long-term overflow using the 

TRANSPORT block of SWMM. The TRANSPORT model was constructed as a 

simplified network consisting of hypothetical overflows and regulator pipes. The 

simplified model was used to project the annual volume presently discharged and the 

percent of the annual volume captured by the proposed storage facility. 

Table 5.5.a-1 demonstrates that the 6400 m3 (1.7 MG) storage facilities provide 

considerable annual capture of WWF compared to that was presently discharged 

without treatment. 

Table 5.5.a-1. Average Annual Overflow Volume from the Pre- and Post-Retrofit 

Facilities (1971 – 1985) 

Average Annual Overflow Volume 

m3 (MG) 

Pre-Retrofit WWTF Overflow 70,400 18.6 

Post-Retrofit (1) 18,000 4.7 

Percent Capture 75% 

Although the ORF was sized on the two-year design storm, the system annually 

captures only 75% of the WWF that is presently discharged. This occurs as a result of 

the regulator chamber modification required to reduce the hydraulic grade line at the 

151
 
regulator and relieve surcharge and street flooding in the vicinity of the regulator 

chamber. Therefore, the ORF will be active more frequently than overflow presently 

occurs from the existing overflow. 

Design Considerations 

The retrofitting of a sanitary sewered collection system to eliminate overflows through 

storage and disinfection required the following design considerations: 

• Hydraulic Loadings 

• Treatment Objectives 

• Operational Requirements 

• Chemical Storage and Feed Requirements 

• Sludge Handling Requirements 

• Operations and Maintenance Requirements 

Hydraulic Loadings - Hydraulic loadings were developed for the ORF to provide 

retention of WWF for subsequent treatment at the WWTF following a wet-weather event 

and to provide disinfection of WWF beyond the 6 ,400 m3 (1.7 MG) ORF capacity. 

Treatment Objectives – The objective of this retrofit is to eliminate overflow from the 

sanitary sewer system up to the two-year design storm and to provide disinfection of 

WWF beyond the two-year design storm level. 

Operational Requirements - The ORF are designed to operate on a continuous basis 

include: 

• WWF retention 

• High-rate disinfection of flow in excess of ORF capacity 

• Chemical Storage and Feed Requirements - The ORF utilize two chemicals: 

• Liquid Sodium Hypochlorite for disinfection 

• Liquid Sodium Bisulfite for dechlorination 

The liquid sodium hypochlorite system is based on commercial-grade liquid sodium 

hypochlorite (15% solution) which contains the equivalent of 150 kg of chlorine/m3

 of 

solution (1.25 lbs of Cl2/gal of solution). Two metering pumps are designed to deliver a 

dosage of 20 mg/L at the peak design flow with one unit as a backup. The liquid 

sodium hypochlorite is stored in a bulk storage tank with an effective working volume of 

11.4 m3

 (3,000 gal). The available chlorine in sodium hypochlorite solution declines 

with age. The anticipated sodium hypochlorite storage time will be 50 days. 

152
 
The liquid sodium bisulfite storage system is based on commercial grade liquid sodium 

bisulfite (38% solution) that contains appropriately 503.2 kg of sodium bisulfite/m3 of 

solution (4.2 lbs of sodium bisulfite/gal of solution). Two metering pumps are designed 

to deliver sodium bisulfite dosages necessary to dechlorinate effluent with 1.5 mg/L of 

sodium bisulfite per mg/L of chlorine residual. Approximately 1.6 kg of sodium bisulfite 

will be used per kg of chlorine residual. A bulk storage tank with an effective working 

volume of 5.7 m3

 (1,500 gal) will be used for sodium bisulfite storage. 

The sodium hypochlorite and sodium bisulfite metering pumps are flow paced. An 

ultrasonic type level sensor is used for effluent flow measurement. 

Sludge Handling Requirements – Accumulated sludge will be returned to the WWTF 

during the pump back of retained WWF and the associated tank flushing following tank 

draining. Tank flushing will be accomplished manually using a 7.6 cm (3 in.) municipal 

water service. 

Operation and Maintenance Requirements - The implementation of the ORF will create 

an O&M cost. The areas of expected O&M burden include: 

• Chemicals for disinfection and dechlorination 

• Power for tank draining and chemical feed systems 

• Water for tank flushing. 

The estimated annual O&M costs are provided in Table 5.5.a-2. 

Table 5.5.a-2. ORF O&M Costs 

Annual Chemical Costs Cost 

Sodium Hypochlorite $ 8,200 

Sodium Bisulfite $ 3,800 

O&M Labor $10,000 

TOTAL COST $22,000 

Project Costs
 

The estimated construction costs for the ORF are provided in Table 5.5.a-3.
 

153
 
Table 5.5.a-3. Estimated ORF Construction Costs 

Installation 

Item Description Quantity Unit Unit Cost Factor Cost 

Site Work 

Common Excavation 1,529 3 m $26 1.0 $40,000 

Rock Excavation-Blasting Method 574 3 m $131 1.0 $75,000 

Select fill 459 3 m $29 1.0 $13,200 

Sheeting 743 2 m $377 1.0 $280,000 

Site Dewatering 1 Ea $20,000 1.0 $20,000 

New Pavement & Restoration 785 m 2 $24 1.0 $18,800 

Fencing Restoration 61 m $98 1.0 $6,000 

General Site & Bldg Restoration 1 Ea $12,000 1.0 $12,000 

Subtotal $465,000 

Structural & Architectural 3 

CIP Concrete-Base 214 m $523 1.0 $112,000 

CIP Concrete-Walls 191 3 m $1,046 1.0 $200,000 

CIP Concrete-Roof 199 m 3 $1,177 1.0 $234,000 

Control Building, Brick & Block 63 2 m $1,507 1.0 $95,200 

Subtotal $641,200 

Mechanical 

Chemical Metering Pumps 2 Ea $3,500 1.5 $10,500 

Submersible Mixers-1.1 kW 5 Ea $3,700 1.7 $31,450 

Oder Control Blowers-11.2 kW 1 Ea $4,500 1.7 $7,650 

ORF Tank Dewatering Pump-10.2 cm 1 Ea $3,500 1.7 $5,950 

Disinfection Tank Dewatering Pump-5.1 cm 1 Ea $2,200 1.7 $3,740 

Back Flow Preventer-7.6 cm 1 Ea $2,000 1.5 $3,000 

Backwater Valve-53.3 cm 1 Ea $1,500 1.5 $2,250 

Subtotal $64,540 

Special Construction 

Carbon Absorption Unit 1 Ea $60,000 1.3 $78,000 

Chemical Storage Tank (Chlorination) 1 Ea $3,000 1.5 $4,500 

Chemical Storage Tank (Dechlorination) 1 Ea $1,500 1.5 $2,250 

Flow Meter 1 Ea $10,000 1.5 $15,000 

SCADA Tansmitter 1 Ea $15,000 1.5 $22,500 

Jib Crane 2 Ea $2,000 1.0 $4,000 

Misc. Safety Equipment 1 Ea $5,000 1.0 $5,000 

Subtotal $131,250 

Site Piping & Utilities 

106.7 cm Interceptor Relocation 18.3 m $2,625 1.0 $48,000 

91.4 cm Interceptor Relocation 55 m $1,637 1.0 $90,000 

61 cm ORF Inf/Effl. Sewer 9.1 m $1,150 1.0 $10,500 

45.7 cm Interceptor Relocation 4.6 m $984 1.0 $4,500 

182.9 cm Storm Sewer Relocation 67 m $1,970 1.0 $132,000 

Bypass Pumping Interceptor 1 Ea $20,000 1.0 $20,000 

Diversion Structure 1 Ea $31,000 1.0 $31,000 

Storm Sewer Junction Manhole 1 Ea $28,000 1.0 $28,000 

Special Manhloe for Sewer Crossing 1 Ea $34,000 1.0 $34,000 

10.2 cm Force Main 6 m $328 1.0 $2,000 

7.6 cm Facility Water Service 67 m $263 1.0 $17,600 

Wet Tap for Water Main Service 1 Ea $3,000 1.0 $3,000 

Temporary Overflow Bypass 9.1 m $656 1.0 $6,000 

Subtotal $426,600 

Construction Cost Subtotal $1,728,590 

Other Contracts 

Electrical
 15% PCT $1,728,590 $259,289 

Heating & Ventilating
 7% PCT $1,728,590 $121,001 

Plumbing
 2% PCT $1,728,590 $34,572 

Subtotal Constuction Costs $2,143,452 

Estimating Contingency 25% PCT $2,143,452 $535,863 

Total Construction Cost $2,679,315 

Engineering & Legal 20% PCT $2,679,315 $535,863 

TOTAL CONSTRUCTION COST $3,220,000 

154 
Conclusions 

Retrofitting SSOs with ORF can provide a cost-effective alternative in lieu of expansion 

of the sanitary collection system to convey the volume of overflow to the treatment plant 

for primary and secondary treatment. This alternative requires the availability of a 

sufficient amount of land to build such a facility. 

155
 
5.6 Retrofitting for Industrial Wastewater Control in a Combined Sewer System 

EPA has identified maximizing flow to the POTW in combined sewer systems as one of 

the nine minimum controls for CSO abatement (EPA 1995). The objective of this 

control is to reduce the magnitude, frequency, and duration of CSOs that flow untreated 

into receiving waters. Industrial wastewater is often characterized by high-strength 

pollutants that can be detrimental to receiving waters. The control of industrial 

wastewater in a combined sewer system can reduce the impact of the CSO discharges 

on receiving waters. A desktop analysis of the Rockland WWTF in Rockland, ME, was 

evaluated to demonstrate the technique. 

5.6.a. Rockland Wastewater Treatment Facility, Rockland, Maine 

The city of Rockland, Maine owns and operates a secondary WWTF that discharges to 

Rockland Harbor, a Class SC water body. The WWTF was originally constructed in 

1978 with upgrades in 1994 and 1998. The Rockland WWTF provides secondary 

treatment using conventional activated sludge. The WWTF was designed to handle a 

monthly-average flow of 11,000 m3/d (2.9 MGD) and an instantaneous-peak flow of 

22,000 m3/d (5.7 MGD). The WWTF serves a CCS. Rockland’s existing collection 

system services most of the residential population, commercial properties, and industrial 

properties within the limits of the city of Rockland. The system also serves the 

neighboring community of Rockport and the Samoset Resort. The system contains 

three types of sewers: 

• combined sanitary and storm 

• separate sanitary 

• storm 

Nine pumping stations exist serving various areas of the collection system. At present 

there are approximately 21 km (13 miles) of gravity sewer ranging from 15 cm (6 in.) to 

76 cm (30 in.) in diameter. Pipe materials include stone and brick (dating back to the 

Civil War era) as well as more recent installations of vitrified clay, asbestos-cement, 

concrete, cast iron, and PVC pipe. According to the 1998 CSO Facilities Plan, there are 

presently two permitted CSOs located within the system. In addition, five other CSOs 

were located during the Facility Planning effort that began in 1993. In an average year, 

187,000 m3

 (49.3 MG) are discharged from these seven CSOs. 

A majority of the wastewater currently entering the Rockland sewer system is domestic 

sewage from residential and commercial sources. The WWTF also receives a 

significant flow and loading from the FMC Corporation Food Ingredients Division facility. 

This facility manufacturers carrageenan, a seaweed based food additive. The WWTF is 

designed to receive approximately 67% of the organic loading and approximately 22% 

of the DWF from the FMC Facility. Annual average wastewater flows to the WWTF are 

7,570 m3

/d (2.0 MGD) and average daily BOD5 loadings are 1,565 kg/d (3,445 lbs/d). 

Prior to implementation of the CSO Abatement Program, the WWTF used the following 

unit processes to provide preliminary, primary and secondary treatment: 

156
 
• Three raw sewage pumps with a capacity of 10 m3 (2,600 gal) per min each, 

• One 92 cm (36 in.) comminutor with a manual bypass screen, 

•	 Two primary clarifiers (each at 14 m [45 ft] with a total surface area of 

approximately 295 m2

 [3,180 ft2]). Primary sludge is conveyed to two cyclone 

classifiers to remove grit prior to thickening, 

•	 Six mechanically aerated activated sludge aeration tanks with a total volume 

of 6,590 m3

 (1.74 MG), 

•	 Two secondary clarifiers (each at 17 m [55 ft] with a total surface area of 

approximately 440 m2 [4,750 ft2 ]), 

• A chlorine contact tank with a volume of approximately 345 m3 (91,000 gal), 

•	 A dechlorination chamber with a total volume of approximately 15 m3 (4,000 

gal). 

Figure 5.6.a-1 shows the existing site layout prior to implementing the CSO Abatement 

Program. Figure 5.6.a-2 shows a schematic of the existing WWTF prior to retrofitting 

with a description of the numbered overflow points. When influent peak flows would 

reach 18,930 m3 (5 MGD), the WWTF influent pumps would be throttled and excess 

flow bypassed to the outfall downstream of the chlorine contact tank resulting in 

untreated, undisinfected discharges. 

These overflows are discharged to Rockland Harbor via permitted CSO #002 (located in 

the wetwell of the WWTF influent pump station Figure 5.6.a-2). The wetwell receives 

the FMC Industrial Wastewater plus WWTF sidestream flows from the primary sludge 

gravity thickener overflow, waste activated sludge DAF thickener overflow, and sludge 

dewatering system vacuum filter filtrate. The industrial wastewater and WWTF 

sidestream flows are mixed with the combined sewage upstream of the influent pump 

station wetwell. Consequently, these high-strength wastes are included in the overflow 

discharged via CSO #002. 

In March 1998, a CSO Facilities Plan was completed. The CSO Facilities Plan was 

approved by the State of Maine Department of Environmental Protection (DEP) and 

EPA. The approved plan recommended a three-phase program to address CSO 

discharges. 

Phase I of the CSO Abatement Program included rehabilitation of the Lindsey Brook 

Sewer. This sewer is located within the streambed for Lindsey Brook and has been 

determined to be a considerable source of I/I during wet weather or high groundwater 

periods. The Lindsey Brook Sewer is also suspected to be a source of exfiltration 

during periods of dry weather. 

157
 
Figure 5.6.a-1. Existing Site Layout Prior to Retrofit 

158
 
Figure 5.6.a-2. Process Flow Schematic Prior to Retrofit 

159
 
Phase II of the CSO Abatement Program included upgrading the WWTF to handle 

additional WWF. Phase II also includes new consolidation conduit to intercept 

overflows from the second and third largest CSO’s (#004 and #005) that are located 

approximately 92 m (300 ft) upstream of the WWTF on Main Street (Route 1) in 

downtown Rockland. Phase II also includes a new force main to separate industrial 

wastewater produced at the FMC Food Ingredients Division Facility from the City 

interceptor sewer and provide a direct connection to the Rockland WWTF. Finally, 

Phase II will redirect WWTF sidestream flows to the primary and secondary treatment 

units downstream of the influent pumping station. 

Phase III of the CSO Abatement Program includes selected sewer separation in the 

south portion of the Rockland CCS to address the remaining CSO discharges. 

This case study will focus on Phase II Improvements at the Rockland WWTF. Phase II 

Improvements include: 

•	 A new 76 cm (30 in) diameter consolidation conduit from CSO #004 and CSO 

#005 to the WWTF. 

• A new flow diversion structure (SS#1) at the WWTF. 

•	 Conversion of the existing headworks into a submersible pumping station to 

handle DWF and WWF. Two dry weather pumps are provided. Each dry 

weather pump has a capacity of 15 m3 (4,000 gal) per min. Three wet 

weather pumps are provided. Each wet weather pump has a capacity of 30 

m3

 (7,800 gal) per min. 

•	 Construction of a new headworks with mechanically cleaned bar screen, 

screenings compactor, manually cleaned bar rack, grit chamber, recessed 

impeller grit pump, cyclone grit classifier, and 31 cm (12 in.) parshall flume. 

•	 Construction of two new upflow style flow distribution structures for the 

existing primary and secondary clarifiers to improve hydraulic flow splitting. 

•	 Construction of a new vortex separator to provide high-rate primary treatment 

for projected peak wet weather CSO flows of 127,560 m3

/d (33.7 MGD). 

•	 Construction of a new CSO Disinfection and Dechlorination Structure to 

provide high-rate disinfection and dechlorination for projected peak WWFs of 

114,690 m3

/d (30.3 MGD); 127,560 m3/d (33.7 MGD) peak WWF minus 

12,870 m3/d (3.4 MGD) vortex underflow. 

•	 Constructions of a new 20 cm (8 in.) diameter force main for industrial 

wastewater. This force main would convey industrial wastewater to the 

existing primary clarifiers, existing aeration tanks, or a new 1,100 m3

 (0.29 

MG) equalization tank. The equalization tank was created by refurbishing 

one (1) unused existing aeration tank. The equalization tank utilizes diffused 

aeration for mixing. The equalization tank contains two submersible pumps 

(1 primary and 1 standby). Each pump has a variable speed drive to allow 

output to be adjusted. 

160
 
•	 Construction of new plant drain piping to eliminate discharge of WWTF 

sidestreams to the influent sewer. The gravity thickeners for primary sludge 

are eliminated by replacing the recessed impeller primary sludge pumps with 

new positive displacement units designed to remove thickened sludge directly 

from the primary clarifiers. The waste activated sludge filtrate drain was 

redirected to the existing aeration tank inlet channel. The sludge dewatering 

filtrate drain was redirected to the primary clarifier influent. 

Figure 5.6.a-3 shows a site layout of the Rockland WWTF following construction of the 

Phase II CSO Abatement Program. Figure 5.6.a-4 shows a process flow schematic of 

the Rockland WWTF following Phase II of the CSO Abatement program. These 

improvements will allow high-rate primary treatment and disinfection for peak wetweather CSO flows at 128,000 m3

/d (33.7 MGD). This provides treatment for CSO 

#002, #004, and #005 that represent approximately 90% of the total annual overflow 

volume from the Rockland CCS. 

Stormwater Management Model Results 

The EPA SWMM Version 4 was used in the facility plan to predict overflow rates and 

volumes at all of Rockland’s CSOs. The SWMM is a relatively complex mathematical 

model. To model the various physical processes associated with stormwater runoff and 

CSO discharge, a large amount of input data is required. Drainage basin characteristics 

(such as size, average ground slope, width, percent imperviousness, roughness, 

surface storage, soil infiltration and gutter or pipe flow) must be supplied to the model. 

The RUNOFF and EXTRAN blocks of SWMM were used in this study. The model was 

constructed using records of the physical system: pipe sizes, slopes, materials, lengths 

and elevations. Each CSO diversion structure was modeled. The existing Park Street 

Pumping Station was also modeled. Then drainage area characteristics were modeled 

for such factors as total acres, percent imperviousness, slope and dimensions. Once 

the SWMM model was mathematically constructed, it was calibrated using the extensive 

rainfall and CSO quantity data gathered during 1993, 1994 and 1995. 

Unlike the calibration process, where actual "real-time" rainfall data are used, SWMM 

projections are made using two types of rainfall data: 

•	 Synthetic rainfall hyetographs (or designs storms) for determining peak rates 

of overflow. 

•	 Long-term rainfall records from NOAA for determining statistically based 

overflow volumes. 

The former data are used in the evaluation of pipes and conduits, pumps and CSO 

treatment facilities where peak rates of discharge are used as the basis of design. The 

latter data are used to design storage facilities and to test the long-term effectiveness of 

overall CSO abatement programs. 

161
 
Figure 5.6.a-3. Site Layout Following Retrofit 

162
 
Figure 5.6.a-4. Process Flow Schematic Following Retrofit 

163
 
For the city of Rockland CCS, the predictions were made using design storms with 

return frequencies of 0.5, 1, 2 and 5 years. A two-hour storm duration was selected for 

the design to reflect an average of time of concentration of the Rockland system. The 

SWMM Modeling identified several important hydraulic characteristics of Rockland’s 

CCS: 

•	 CSO #005 has the highest projected peak overflow rate, CSO #002 (WWTF 

Bypass) has the second highest peak overflow rate. 

•	 CSO #002 (WWTF Bypass) discharges the greatest overflow volume with the 

North Subsystem overflows (CSO #004 and #005) discharging the second 

greatest volume. 

Table 5.6.a-1 shows the predicted overflow activity in terms of average annual volume 

of overflow per year. These data were developed using the SWMM model and 

continuous simulation. Tables 5.6.a-2 and 5.6.a-3 show the projected peak overflow 

rates for each CSO using the design storms. 

Table 5.6.a-1. SWMM Modeling Summary - Annual Overflow Volumes 

Estimated Estimated 

Annual Annual 

Drainage Area 

Description 

CSO 

ID Number 

CSO Discharge 

(m3

/yr.)(1) 

CSO Discharge 

(MG/yr.)(1) 

WWTF Bypass 002 110,900 29.3 

North/Lindsey/Downtown 004 and 005 54,000 14.4 

003, 006, 007 

South and 008 21,200 5.6 

TOTAL 186,100 49.3 

(1)Using long-term simulation based upon 21 years of hourly rainfall data form the 

NWS station in Portland, ME for the years from 1971 to 1991. 

Table 5.6.a-2. SWMM Modeling Summary - Projected Peak Flows 

Estimated Peak Flow (m3/d) 

CSO 

ID Number 

0.5 Year 

Storm(1) 

1 year 

Storm(1) 

2 year 

Storm(1) 

2 year 

Storm(1) 

002 23,850 31,800 33,000 34,100 

003 5,300 5,300 5,300 5,300 

004 0.0 10,100 23,100 37,100 

005 47,700 53,800 54,900 56,000 

006 380 6,440 14,000 21,200 

007 6,440 6,440 6,440 6,440 

008 1,900 3,400 5,300 7,600 

(1) Using 2 h duration. 

164
 
Table 5.6.a-3. SWMM Modeling Summary - Projected Peak Flows 

Estimated Peak Flow (MGD) 

CSO 0.5 Year 1 year 2 year 2 year 

ID Number Storm(1) Storm(1) Storm(1) Storm(1) 

002 6.3 

003 1.4 

004 0.0 

005 12.6 

006 0.1 

007 1.7 

008 0.5 

(1)Using 2 h duration. 

8.4 8.7 9.0 

1.4 1.4 1.4 

2.9 6.1 9.8 

14.2 14.5 14.8 

1.7 3.7 5.6 

1.7 1.7 1.7 

0.9 1.4 2.0 

Design Considerations 

The following is a list of key design considerations for the Rockland WWTF retrofit: 

• Impacts of High Strength Wastes 

• Hydraulic Loadings 

• Treatment Objectives 

• Operational Requirements 

• Chemical Storage and Feed Requirements 

• Sludge Handling Requirements 

• Operations and Maintenance Requirements 

• Construction Sequencing and Site Constraints 

Each of these considerations is described in this section. 

Impacts of High-Strength Wastes. One of the nine minimum CSO controls is the review 

and modification of pretreatment programs to assure that CSO impacts are minimized. 

The elimination of the industrial wastewater connection from FMC Food Ingredients 

Division to the city of Rockland CCS, and construction of a separate force main to the 

WWTF will satisfy this objective. 

The industrial wastewater design loading to the Rockland WWTF is approximately 2,990 

kg (6,583 lbs) per day. This is approximately 67% of the total organic loading of 4480 

kg (9,871 lbs) that the WWTF receives daily. Following the implementation of Phase II 

improvements, this industrial wastewater can be directly discharged to three locations at 

the Rockland WWTF: 

• Existing Primary Clarifier Influent at New Flow Distribution Structure No. 1 

• Existing Aeration Tank Inlet Channel 

• Equalization Tank 

165
 
The industrial wastewater enters the WWTF via a 20 cm (8 in.) diameter force main 

connected to a new pre-cast concrete valve vault (SS #2). SS #2 contains valves to 

direct flow to the three discharge locations. A 20 cm (8 in.) magnetic flow meter 

measures flow. If the flow is directed to the equalization tank, then it is mixed using a 

fine bubble diffused air system. The design airflow is 30 scfm per thousand cubic feet. 

The air blower has a variable frequency drive allowing adjustments to the airflow. The 

WWTF staff is able to mix the equalization basin at a lower rate, if desired. A total of 

324 membrane diffusers each with a maximum airflow of 3.5 scfm are provided in the 

equalization basin. 

Two submersible pumps (1 primary and 1 standby) convey industrial wastewater from 

the equalization basin to the WWTF. Each pump has a rated capacity of 2.3 m3

 (600 

gal) per min at a total dynamic head of 4.5 m (14 ft). The pumps are driven with 3.7 kW 

(5 HP) motors controlled with variable frequency drives that allow the WWTF staff to 

adjust the flow of industrial wastewater to the process. The discharge from the 

submersible pumps can be directed to two locations: 

• Existing Primary Clarifiers via new Flow Distribution Structure No. 1 

• Existing Aeration Tank Inlet Channel 

A 15 cm (6 in.) diameter magnetic flow meter is provided on the submersible pump 

discharge allowing the rate of industrial wastewater flow to be monitored by the WWTF 

staff. 

Other high-strength wastes eliminated from the CSO discharge with the Phase II 

Improvements are the sidestream flows from the sludge processing system. The 

sidestream flows represent a relatively small portion of the daily load. Approximately 

127 kg (279 lbs) per day are discharged (approximately 3% of the overall design 

organic loading of 4480 kg (9,871 lbs) per day). However, these loadings are 

concentrated within a few hours during the day, which magnifies their effects. 

The sludge thickener filtrate is produced in a 3.5-4 h period and represents 

approximately 8% of the hourly organic loading while the sludge de-watering filtrate is 

produced in a 3.5-4 h period and represents approximately 12% of the hourly organic 

loading. Overall, the sidestreams represent approximately 20% of the hourly organic 

loading at the WWTF when the units are in service. The sludge thickening sidestream 

was redirected to the existing Aeration Tank Inlet Channel while the sludge dewatering 

sidestream was redirected to the existing Primary Clarifier influent at new Flow 

Distribution Structure No. 1 with the Phase II Improvements. 

Hydraulic Loadings. Pollutant and hydraulic loadings were developed for the wetweather treatment units at the Rockland WWTF during the Facilities Planning and 

Design stages. The primary objective of the wet-weather treatment units is to remove 

floatable and settleable solids to promote effective disinfection. The primary objective of 

the industrial wastewater equalization basin is to remove high-strength industrial 

wastewater from the CCS to allow full secondary treatment of these wastes at all times. 

166
 
Table 5.6.a-4 shows the design criteria for the wet-weather treatment processes at the 

Rockland WWTF. 

Table 5.6.a-4. Wet-Weather Treatment Design Criteria 

Description and Criteria Phase II Design Value 

A. Flows and Loadings 

WWTF Peak Monthly Avg. Flow 

WWTF Peak Daily Flow 

WWTF Instantaneous Maximum 

Total Annual CSO Volume 

CSO Annual Volume at WWTF 

10,100 m3

/d 2.90 MGD 

12,500 m3

/d 3.30 MGD 

21,600 m3

/d 5.70 MGD 

186,600 m3 49.3 MG 

188,100 m3 43.7 MG 

CSO North/West Subsystem Peak (5 yr., 2 h) 127,550 m3

/d 33.7 MGD 

CSO North/West Subsystem Peak (2 yr., 2 h) 110,900 m3

/d 29.3 MGD 

CSO North/West Subsystem Peak (1 yr. 2 h) 96,500 m3

/d 25.5 MGD 

CSO North/West Subsystem Peak (1/2 yr. 2 h) 71,500 m3

/d 18.9 MGD 

B.	 Influent Pumping - Dry Weather 

No. of Pumps (each) 2
 

Type 

Flow 

Non-clog Submersible
 

15 m3

/min 4,000 gpm
 

TDH 9 m 30 ft
 

Power (each) 37.3 kW 50 HP 

C.	 Screenings Handling System 

Type Mechanical Screen 

No. of units (each) 1 

Channel Width 91.0 cm 36 in. 

Bar Screen Clear Spacing 2.5 cm 1.0 in. 

Bar Size 6.5 x 1.0 cm 2-1/2 x 3/8 in. 

Headloss, half clogged 22.9 cm 9.0 in. 

D.	 Grit Removal 

Type Pista
 

No. of Units (each) 1
 

Diameter 3.7 m 12.0 ft
 

Hopper Diameter 1.5 m 5.0 ft
 

Type of pump Recessed Impeller
 

No. of pumps (ea.) 

Flow 

1
 

1 m3

/min 260 gpm
 

TDH 9 m 30 ft
 

Horsepower (each) 10
 

Type of De-watering 

Pressure 

Cyclone
 

4,922 kg/m2 7.0 psig 

Screw Diameter 30.5 cm 12.0 in. 

E.	 Wet Weather Pumping 

No. of pumps (each) 3
 

Non-Clog Submersible
 

Type 

167
 
Table 5.6.a-4. Wet-Weather Treatment Design Criteria, continued 

Description and Criteria Phase II Design Value 

F.	 

G.	 

Flow each
 

TDH
 

Power each
 

Maximum Speed (rpm)
 

Minimum Efficiency (%)
 

Vortex Separator 

No. of Units (each) 

Type 

Diameter each 

Sidewater Depth each 

5 year Hydraulic Loading 

2 year Hydraulic Loading 

1 year Hydraulic Loading 

1⁄2 year Hydraulic Loading
 

Influent Pipe Size
 

Effluent Pipe Size
 

Underflow Pipe Size
 

Underflow Rate, Minimum
 

Underflow Rate, Maximum
 

No. of Underflow pumps
 

Type of pump
 

Pump Flow (each)
 

TDH
 

Motor power each
 

Minimum Efficiency (%)
 

Maximum Pump Speed (rpm)
 

CSO Disinfection And Dechlorination 

No. of Units (each) 

Volume 

Length to Width Ratio 

Chlorine Detention Time – 5 yr. 

Q = 114,685.5 m3

/d (30.3 MGD) (min) 

Chlorine Detention Time – 2 yr. 

Q = 98,031.5 m3

/d (25.9 MGD) (min) 

Chlorine Detention Time – 1 yr. 

Q = 83,648.5 m3

/d (22.1 MGD) (min) 

Chlorine Detention Time – 1⁄2 yr. 

Q = 58,667.5 m3/d (15.5 MGD) (min) 

Dechlorination Tank Volume 

Dechlorination Detention Time – 5 yr. 

Q = 114,685.5 m3

/d (30.3 MGD) (s) 

No. of NaHSO3 pumps (each) 

Capacity of NaHSO3 Pumps (each) 

No. of Sodium Bisulfite Pumps (each) 

Capacity of NaHSO3 Pump (each) 

30 m3

/min 7,800 gpm
 

8.5 m 28 ft
 

74.6 kW 100 HP
 

720
 

76
 

1
 

EPA
 

10.7 m
 35 ft
 

2.7 m
 8.75 ft
 

1,430 m3/d/m2
 35,000 gpd/ft2
 

1,240 m3/d/m2
 30,400 gpd/ft2
 

1,080 m3/d/m2
 26,500 gpd/ft2
 

780 m3/d/m2
 19,600 gpd/ft2
 

107 cm
 42 in.
 

107 cm
 42 in.
 

35.5 cm
 14 in.
 

6435 m3

/d
 1.7 MGD
 

12,870 m3

/d
 3.4 MGD
 

2
 

Self Priming Non-Clog
 

4.5 m3

/min 1,170 gpm
 

4.3 m 14 ft
 

7.5 kW 10 HP
 

55
 

860
 

1 

400 m3 105,300 gal 

40:1 

5.0 

5.8 

6.7 

9.2 

19 m3 5,100 gal 

15
 

2+1 standby
 

0.4 m3

/h 105 gph
 

1+1 standby
 

0.1 m3

/h 25 gph
 

168
 
Treatment Objectives. Sampling and monitoring performed during the development of 

the CSO Abatement Program indicated that Rockland Harbor does not meet current 

water quality criteria for enterococcus bacteria during wet weather. Currently, the State 

of Maine allows a maximum geometric mean enterococcus concentration of 14 colonies 

per 100 ml and a daily maximum enterococcus concentration of 94 per 100 ml to 

comply with Standards established for Class SC waters. The primary focus of the 

phase II Improvements are: 

•	 Increasing the quantity of CSO conveyed to the WWTF for treatment and 

disinfection. 

•	 Segregating high-strength industrial wastewater and WWTF sidestreams from 

the overflow at the influent pump station. 

•	 Providing high-rate primary treatment to remove floatable and settleable 

solids from the CSO to facilitate disinfection. 

The CSO Abatement Program recommended a plan to treat CSO discharges to meet 

the current maximum daily water quality limits. The high-rate primary treatment and 

disinfection processes were selected as the most economical method to meet the 

treatment objectives to meet water quality objectives for this project. 

Operational Requirements. The Phase II Improvements to the Rockland WWTF which 

are designed to operate on a continuous basis include: 

• Dry Weather Pumping Station 

• Industrial Wastewater Equalization 

• Mechanical Screening 

• Grit Removal 

• Primary Treatment 

The wet-weather pumping system, high-rate primary treatment (vortex separator) highrate disinfection, and dechlorination systems are intended to operate in an event-based 

mode during periods of excessive WWF. Under normal dry weather conditions flow 

from the north and south CCS enter a new diversion structure (SS #1) upstream of the 

influent pumping station. SS #1 contains a side overflow weir separating the overflow 

from the normal sewage flow through the structure. As sewage flows increase during a 

wet-weather event, the new submersible dry weather pumps will be throttled to control 

flow to the secondary WWTF. As water levels rise in the dry-weather pump wetwell flow 

will automatically be diverted to the overflow chamber. These overflows are conveyed 

to the wetwell of the wet-weather pumping system. When levels in the wet-weather 

wetwell rise, the wet-weather pumps will automatically start and convey flow to the new 

vortex separator via a 1 .0 m (42 in.) force main. Once flow through the vortex separator 

has been established, a level switch located at the inlet to the high-rate CSO 

Disinfection and Dechlorination Structure will start the vortex underflow pumping 

system. The vortex underflow pumps are rated to deliver 4.50 m3 (1,170 gal) per min at 

169
 
4.5 m (14 ft) total dynamic head. Each pump has a 7.5 kW (10 HP) electric motor with a 

variable frequency drive allowing the WWTF staff to adjust the rate of vortex underflow. 

The level switch located at the inlet of the CSO Disinfection and Dechlorination 

Structure will also start the chemical metering pumps (sodium hypochlorite for 

disinfection and sodium bisulfite for dechlorination) and the tank mixers. 

An ultrasonic flow meter is provided at the 160 cm (64 in.) broad crested rectangular 

weir at the inlet of the CSO Disinfection and Dechlorination Structure. The flow meter 

records CSO flows, and provides a 4-20 mA signal to control the speed of the sodium 

hypochlorite and sodium bisulfite chemical metering pumps. Flow pacing of the 

chemical feed rate will improve disinfection and dechlorination efficiency, and reduce 

overall chemical consumption. 

At the end of the wet weather event automatic level controls will stop the submersible 

wet-weather pumps. When the influent flow stops, the level in the vortex separator will 

fall as it is drained with the vortex underflow pumping system. The chemical metering 

pumps and tank mixers automatically stop when the level in the CSO Disinfection and 

Dechlorination Structure falls below the pre-selected set point. The WWTF staff would 

then drain the CSO Disinfection and Dechlorination Structure and wash down the 

structures to prepare for the next event. 

Chemical Storage and Feed Requirements. The CSO Abatement facilities will utilize 

two chemicals: 

• Liquid Sodium Hypochlorite for disinfection 

• Liquid Sodium Bisulfite for dechlorination 

The liquid sodium hypochlorite system is designed for commercial-grade liquid sodium 

hypochlorite (15% solution) which contains the equivalent of 150 kg of chlorine/m3

 (1.25 

lbs Cl2/gal) of solution. The metering pumps can deliver a dosage of 25 mg/L at the 

peak design flow with one unit out of service. The bulk storage tank has an effective 

working volume of 4 m3 (1,000 gal). This tank is interconnected with the existing 15 m3 

(4,000 gal) sodium hypochlorite bulk storage tanks used for secondary effluent 

disinfection, giving the WWTF a total of 19 m3 (5,000 gal) of hypochlorite storage. The 

available chlorine in sodium hypochlorite solution declines with age, the anticipated 

sodium hypochlorite storage time will be 52 days at the design flow of 10,980 m3/d (2.9 

MGD). 

The liquid sodium bisulfite storage system is designed for commercial grade liquid 

sodium bisulfite (38% solution) which contains appropriately 500 kg sodium bisulfite/m3 

(4.2 lbs NaHSO3/gal) of solution. The metering pumps can deliver sodium bisulfite 

dosages necessary to dechlorinate effluent with a chlorine residual of 5.0 mg/L. 

Approximately 1.6 kg of sodium bisulfite will be used per kg of chlorine residual. The 

bulk storage tank has an effective working volume of 4 m3

 (1,000 gal). This tank is 

interconnected with the existing 8 m3 (2,000 gal) sodium bisulfite tank used for 

170
 
secondary effluent dechlorination, giving the WWTF a total of 12 m3 (3,000 gal) of 

sodium bisulfite storage. 

Sludge Handling Requirements. Sludge handling requirements increase during periods 

of WWF. The most pronounced increase occurs during the initial period of WWF (first 

flush). 

A solids balance was performed to estimate the quantity of sludge produced during a 

one yr frequency, two hr duration design storm. The daily sludge quantities would 

increase as follows: 

• Estimated Sludge Quantity 2,870 kg (6,326 lbs) per d 

• One-Year Storm Estimated Sludge Quantity 3,640 kg (8,029 lbs) per d 

The sludge quantity is estimated to increase by approximately 27% during a one-year 

storm event. The overall increase can be assimilated if sludge pumping and processing 

schedules are adjusted during the remainder of the week. The following is a summary 

of weekly sludge quantities during a period with a one-year duration, two-hour 

frequency design storm: 

• Initial Estimated Sludge Quantity (kg/week) 20,090 

• Projected Sludge with One-Year Design Storm (kg/week) 20,860 

The estimated weekly sludge quantity would increase by approximately 4% during a 

week with a one-year storm event. 

Operation and Maintenance Requirements. The implementation of the Phase II 

Improvements will result in increased O&M costs. The following list summarizes the 

areas of expected O&M cost increases at the WWTF: 

•	 Labor to perform sampling, laboratory analyses, reporting, wet-weather 

process adjustments, and clean-up following an event. 

• Power for in-plant pumping, sludge processing and chemical feed systems. 

•	 Chemicals for disinfection, dechlorination, and polymer for additional sludge 

quantities. 

• Sludge disposal tipping and transport fees. 

Table 5.6.a-5 summarizes the expected increase in O&M costs at the WWTF resulting 

from the Phase II Improvements. 

171
 
Table 5.6.a-5. Estimated O&M Costs Phase II CSO Improvements 

Estimated Annual 

Description Cost 

Sampling
 $600 

Laboratory Analyses
 2,500 

Report Preparation
 1,200 

Process Control Adjustments
 1,000 

Clean-Up of Wet-Weather Treatment Structures
 4,500 

Power
 700 

Sodium Hypochlorite
 3,100 

Sodium Bisulfite
 1,600 

Polymer for Sludge De-watering
 300 

Sludge Disposal Transport and Tipping Fees
 4,000 

Total Estimated Cost $19,500 

Construction Sequencing and Site Constraints. The existing Rockland WWTF site is 

crowded with little room for new facilities in the area where the influent sewers enter the 

WWTF. Therefore, it was necessary to construct some of the CSO Abatement Facilities 

in locations previously used for other purposes. The following is a listing of special 

sequencing and site considerations related to the CSO Abatement Program at the 

WWTF: 

•	 A storage garage was demolished to facilitate construction of the vortex 

separator. 

•	 There was limited space available for wet-weather pumping. The existing 

preliminary treatment building (which contained the influent pumps) was 

converted into a submersible pumping station for DWF and WWF. The dry 

weather pumps are installed in the existing wetwell. The wet-weather pumps 

are installed in the existing pump room that is converted into a wetwell for the 

submersible wet-weather pumps. The modifications will require temporary 

bypass pumping of flow into the WWTF during construction. 

•	 New junction structures are required on the existing outfall pipe (SS #4 and 

SS #5) to increase the hydraulic capacity. An emergency relief discharging to 

Lermond Cove (adjacent to the WWTF site) will be preserved. 

A substantial completion time of 510 calendar days was specified for the Phase II 

Improvements. 

172
 
Project Costs 

The bids for the Phase II Improvements were opened on July 8, 1998. Costs for the 

CSO related components of the project are listed in Table 5.6.a-6. Costs include 

construction requirements and project services. Project services include design 

engineering, construction engineering, legal, and fiscal expenses. Project services 

costs do not include contingencies, right-of-way expenses, or other miscellaneous 

items. 

Table 5.6.a-6. Project Costs -Phase II CSO Improvements 

Description Cost1 

General Conditions 

Site Work and Yard Piping 

Consolidation Conduit 

Special Structure No. 1 

Dry Weather/Wet Weather Pumping Station Renovation 

Headworks/CSO Disinfection and Dechlorination 

Structure
 

Vortex Separator
 

Special Structure No. 2
 

Equalization Tank Renovation
 

Industrial Wastewater Force Main
 

Flow Distribution Structure No. 1
 

Flow Distribution Structure No. 2
 

Special Structure No. 4
 

Special Structure No. 5
 

Yard Electrical Ductbanks and Lighting
 

Instrumentation and SCADA System
 

$175,000 

375,000 

140,000 

90,000 

180,000 

1,080,000 

240,000 

40,000 

58,000 

22,000 

60,000 

65,000 

60,000 

55,000 

90,000 

70,000 

Total Construction Cost 2,800,000 

Project Services2 600,000 

Total Project Cost 3,400,000 

1

Based on July 8, 1998 Bid Prices

2

Includes Design Engineering, Construction Engineering, Legal, and Fiscal 

Expenses. Does not include contingencies and right-of-way costs. 

Conclusions 

The retrofit of an unused aeration tank and the construction of a separate force main 

from the FMC Corporation Food Ingredients Division facility to the WWTF provides a 

cost-effective alternative for the elimination of a established portion of high-strength 

wastes from CSO discharges at the Rockland WWTF. This retrofit, combined with other 

WWTF improvements directed at CSO abatement, will direct high-strength waste from 

the FMC facility to the dry-weather headworks, eliminating the potential for CSO 

discharge of industrial wastewater. This retrofit was part of an overall CSO Abatement 

Program necessary to address system-wide CSO discharges. 

173
 
5.7	 Bringing Outdated/Abandoned POTWs Back Online as Wet-Weather Flow 

Treatment Facilities 

The wastewater treatment facilities remaining following upgrades or construction of new 

POTWs are often abandoned or demolished. These outdated or abandoned facilities 

could possibly serve a useful function as part of the new treatment facilities depending 

on their condition and location with respect to the new treatment facilities and the 

combined sewer system. These facilities may be utilized as wet-weather storage or 

flow equalization facilities. 

5.7.a. Auburn WWTF, Auburn, New York 

Background 

The city of Auburn Water Pollution Control Plant was initially constructed during the 

1930s and provided primary treatment. In 1970, the original primary treatment facilities 

were upgraded to provide secondary treatment. Since that time, several improvements 

have been completed. 

The WWTF includes preliminary treatment (screening and grit removal), primary 

sedimentation, trickling filtration, final sedimentation, and chlorination. Sludge treatment 

facilities include gravity sludge thickeners, belt filter press for sludge dewatering, and 

incinerators for sludge disposal. The ash from the incineration process is landfilled. 

The Auburn WWTF services the city of Auburn and the surrounding Towns of Aurelius, 

Fleming, Sennett, and Owasco. Portions of the existing sewer system consist of 

combined sewers that carry both sanitary sewage flow and storm sewer flow in the 

same pipe. 

In December 1989, the city of Auburn entered into a consent order agreement with the 

New York State Department of Environmental Conservation (NYSDEC). The consent 

order required the city of Auburn to complete an Infiltration/Inflow Analysis, a CSO 

Analysis, BMP Program, and a Comprehensive Performance Evaluation (CPE) for the 

city of Auburn Water Pollution Control Plant. 

Phase I improvements to the city of Auburn WWTF were performed between 

1993-1995, as summarized below: 

1.	 New preliminary treatment facilities consisting of mechanical bar screens 

housed in a new influent building, aerated grit channels, and new influent flow 

metering. 

2.	 New primary treatment facilities, including new primary settling tanks and 

sludge handling pumps and rehabilitation of the existing- settled sewage 

pumps and controls. 

174
 
3.	 A new ORF utilizing the abandoned primary settling tanks, with chlorination 

(using sodium hypochlorite) and dechlorination (using sodium bisulfite) facilities 

and flow metering equipment for the ORF overflow. 

4.	 Trickling filter rehabilitation, including replacement of rotary distribution arms, 

rehabilitation of existing recirculation pumps and controls, and installation of 

recirculation flow metering equipment. 

5.	 New septage receiving facilities, including an enclosed septage receiving 

station, mechanical septage screening, underground septage storage tank, and 

septage handling equipment. 

6.	 Odor control facilities to treat ventilated air from influent building, screenings 

and grit loading area, septage receiving station and septage storage tank, and 

belt press room of the existing administration building. 

Construction costs for the Phase I improvements were estimated to be $9,600,000. 

Figure 5.7.a-1 shows the WWTF flow schematic prior to construction of the Phase I 

improvements 

The following summarizes the retrofit of the abandoned primary treatment tanks for use 

as overflow retention facilities for the treatment of WWF in excess of the treatment plant 

capacity. 

Previous Studies 

As a requirement of the consent order with New York State, the city of Auburn has 

completed studies on I/I and CSO abatement. 

Infiltration and Inflow Study. In January 1991, the city of Auburn completed an I/I study 

of the wastewater collection system. The purpose of this study was to develop a 

recommended plan of corrective action to eliminate excessive I/I within the city's 

sanitary sewer system. The study recommended the removal of sewer system 

conveyance restrictions and the elimination of direct inflow sources. The results of the 

I/I study indicated that infiltration was not excessive and that it was more cost effective 

to transport and treat, than to eliminate. The study also suggested that approximately 

31,420 m3

/d (8.3 MGD) of peak inflow could cost effectively be removed. 

CSO Abatement. In July 1993, the city of Auburn completed a CSO abatement 

study/comprehensive water collection system plan. The CSO abatement study included 

the evaluation of alternatives for the elimination, storage, and/or treatment of CSOs that 

discharge to the Owasco Outlet within the city of Auburn. The CSO study identified the 

alternative of centralized high rate treatment as the most cost effective and 

environmentally sound alternative for the city of Auburn's collection system. This 

alternative includes the construction of a new interceptor to increase the capacity of the 

existing North Owasco and South Interceptors, and the construction of a high rate 

treatment facility and diversion structure at the existing WWTF. The CSO study 

recommended the implementation of the North Interceptor improvements as described 

175
 
5.7.a-1 Existing Wastewater Treatment Flow Diagram 

176
 
in the North Interceptor Replacement Sewer Basis of Design Report dated April 1993. 

Also included in the CSO study was an estimate for the future (year 2010) total peak 

flow to the WWTF of 191,000 m3

/d (50.46 MGD). Of this total flow, 44,970 m3/d (11.88 

MGD) was estimated for total base flow plus infiltration. 

Comprehensive Performance Evaluation. In June 1992, an interim CPE report for the 

city of Auburn was completed. This interim report developed the conceptual WWTF 

improvements required to meet the anticipated reclassification of Owasco Outlet and 

revised effluent limits. 

Project Approach 

The CSO report recommended the construction of peak flow storage facility on the 

North Interceptor and the construction of parallel interceptors to convey peak flows to 

the WWTF. This report had also recommended the construction of new overflow 

facilities, such as a swirl concentrator, to treat flows above 78,400 m3/d (20.7 MGD). 

These facilities were completed and operation began in July 1996. 

In lieu of the swirl concentrators to treat peak flows at the WWTF, an ORF was 

evaluated. The ORF would treat flows above the 95,000 m3/d (25 MGD) WWTF 

capacity and be constructed from the abandoned primary settling tanks, which became 

available after the treatment plant was upgraded. The reuse of this tankage represented 

a significant cost saving to the city of Auburn. 

Overflow Retention Facilities 

The abandoned primary settling tanks were rehabilitated to function as an ORF. The 

basis of design for the ORF is summarized in Table 5.7.a-1 as described below. 

Table 5.7.a-1. Basis of Design of the Overflow Retention Facilities 

Overflow Retention Tanks Peak hourly flow 96,500 m3/d (25.5 MGD) 

Number of Units 5 (abandoned primary settling tanks)
 

Total Surface Area 782 m2 (8,416 ft2)
 

Maximum SOR 123 m3

/d/m2 (3,025 gpd/ft2)
 

Equipment
 

Overflow Return Pumps
 

Number of Units 

Type Centrifugal
 

Return Duration 24 h
 

Overflow Influent Metering
 

Type Open Channel 

Flow Element Weir 

Flow Meter Ultrasonic 

Range 0—11,355 m3

/d ( 0—3 MGD) 

177
 

4
 
Table 5.7.a-1. Basis of Design of the Overflow Retention Facilities, continued 

High-Rate Disinfection 

Number of Units 

Contact Tank Volume 

Detention Time 

Tank Dimensions (existing) 

(Width x Length x Side Water Depth) 

Mixing Energy 

Chlorination Chemical Feed Equipment 

Type 

Dosage 

Dechlorination 

1 (abandoned primary settling tank #6) 

318 m3

 (11,220 ft3) 

4.7 min at peak hourly flowrate 

5.9 m x 20.9 m x 2.6 m 

(19.5 ft x 68.5 ft x 8.4 ft) 

22.4 kW (30 HP) 

Liquid Sodium Hypochlorite 

25 mg/l 

Detention Time 12 s at peak hourly flow 

Contact Tank Volume 13 m3

 (458 ft3) 

Tank Dimensions 1.5 m x 3.6 m x 2.4 m 

(Width x Length x Side Water Depth) (5 ft x 11.85 ft x 7.74 ft) 

Equipment
 

Dechlorination Chemical Feed
 

Type
 Liquid Sodium Bisulfite 

Dosage per mg/l of chlorine residual
 1.5 mg/l 

Mixers
 

Number of Units
 1 per tank 

Type
 Constant Speed 

Tank Flushing System
 

Source Chlorinated Plant Effluent 

Overflow Effluent Metering 

Type
 Open Channel
 

Flow Element
 Open Channel
 

Flow Element
 Weir
 

Flow Meter
 Ultrasonic
 

Range
 0—113,550 m3

/d (30 MGD)
 

Influent flows in excess of the 95,000 m3/d (25 MGD) WWTF capacity are directed to 

the retention facility through a controlled diversion structure. The maximum SOR 

provided by using five of the six abandoned primary settling tanks as an overflow 

retention tank(s), is 123 m3

/d/m2 (3,025 gal/d/ft2) at the estimated peak hourly flow of 

96,500 m3

/d (25.5 MGD). Overflow events producing less than approximately 4,100 m3 

(1.083 MG), which includes volume disinfection contact chamber, are retained and 

returned to the treatment process when influent flows subside. Overflow events 

producing in excess of the available storage volume are disinfected using chlorine and 

178
 
dechlorinated before being discharged to the Owasco Lake Outlet. The retained 

volume is returned to the aerated grit chambers utilizing four overflow return pumps. 

Figure 5.7.a-2 shows the flow schematic of the Auburn WWTF following construction of 

the ORF. 

The overflow retention facilities include the following: 

Modifications to the Abandoned Primary Settling Tanks. The primary sludge longitudinal 

and cross collectors were removed to use the abandoned primary settling tanks as an 

ORF. Deteriorated concrete surfaces were repaired and weirs were installed to allow 

flow between tanks. A high volume tank flushing system was provided that uses 

chlorinated plant effluent to wash the tank walls and bottom following an overflow event. 

Overflow Return. The old primary sludge pump was replaced with overflow return 

pumps. The new pumps are suitable for pumping grit and primary sludge. In order to 

prevent odors, the entire overflow retention tank volume is returned to the treatment 

process within 24 h after the occurrence of an overflow event provided influent flow 

rates have subsided. 

Disinfection. Overflow events producing stormwater in excess of the available storage 

volume are disinfected using a high-rate disinfection system. The system consists of a 

contact tank sized for a 4 min detention time at peak hourly flows of 96,520 m3/d (25.5 

MGD) and requires the use of mechanical mixers to induce energy to provide effective 

disinfection. The mixing velocity gradient and chlorine dosage were calculated on the 

basis of estimated influent and effluent fecal coliform levels for CSOs, detention time, 

concentration of TKN, BOD, and an estimated value for pH. The high-rate disinfection 

system was sized by estimating disinfectant dosage and mixing energy required to 

assure the inactivation of total coliform. Based on the optimum velocity gradient 

calculated, horsepower requirements for mixers were determined. From this analysis, it 

was determined that an optimum chlorine dosage of 25 mg/l and approximately 22.4 kW 

(30 HP) of total mixing energy are required to insure proper disinfection. 

The abandoned Primary Settling Tank #6 was modified for use as the high-rate 

disinfection contact tank. The chlorination facilities are housed in a new structure 

constructed over a portion of the new overflow retention tank. This structure houses the 

chlorination and dechlorination equipment. The chlorination equipment was designed to 

be automatically flow-paced according to the overflow flow rate. 

Dechlorination. Dechlorination facilities were installed to reduce the chlorine residual 

before discharge to the Owasco Outlet. The dechlorination facilities are metered 

automatically as a function of overflow flow rate and will be designed to minimize 

chlorine residual. The dechlorination tank contains a high-rate mixer designed for a 

detention time of 30 s at peak hourly flows of 96,500 m3

/d (25.5 MGD). This tank was 

constructed within a portion of abandoned Primary Settling Tank #6. 

Overflow Metering. Overflows discharged to the Owasco Lake Outlet are metered after 

being discharged from the dechlorination tank through an open channel-type metering 

device located in a separate concrete structure. 

179
 
Figure 5.7.a-2. Conceptual Wastewater and Sludge Treatment Flow Diagram 

180
 
Stormwater Management Model Results 

The EPA SWMM Version 4 was used to predict the annual capture of WWF that 

previously discharged at the WWTF and is now treated through the ORF. A simplified 

model of the WWTF drainage area was used to determine the impact of pre- and postretrofit. This analysis is based on 21 years of hourly rainfall from the NWS Station at 

Hancock Field in Syracuse, NY. Table 5.7.a-2 demonstrates that use of the ORF 

resulted in a considerable annual capture of WWF that was previously discharged 

without treatment. 

Table 5.7.a-2. Average Annual Overflow Volume from the Pre- and Post-Retrofit 

Facilities (1971 – 1991) 

Average Annual Average Annual 

Overflow Volume Overflow Volume 

Condition (m3

) (MG) 

Pre-Retrofit WWTF Overflow 29,000 7.65 

Post-Retrofit Overflow(1) 7,000 1.85 

Percent Capture 76% 

(1)Overflows receive preliminary treatment and disinfection prior to discharge 

The ORF annually captures 22,000 m3 (5.8 MG) of wet-weather overflow at the WWTF 

that is returned to the WWTF influent for treatment following storm events. 

Design Considerations 

The retrofitting of the abandoned primary settling tank to provide enhanced wet-weather 

treatment through storage and disinfection required careful consideration of several 

factors. The following is a list of key design considerations for this retrofit: 

• Pollutant and Hydraulic Loadings 

• Treatment Objectives 

• Operational Requirements 

• Chemical Storage and Feed Requirements 

• Sludge Handling Requirements 

• Operations and Maintenance Requirements 

Each of these considerations will be described in this section. 

Pollutant and Hydraulic Loadings. Pollutant and hydraulic loadings were developed for 

the ORF during the Facilities Planning stage. These values were refined and updated 

as the project progressed through the preliminary engineering and detailed design 

181
 
stages. The primary objective of the ORF is to provide preliminary treatment and 

disinfection of WWF beyond the 95,000 m3/d (25 MGD) WWTF capacity. The basis of 

design for the ORF is shown on Table 5.7.a-1. 

Treatment Objectives. The rationale for this retrofit was meeting the revised effluent 

limits from the WWTF in the Owasco Lake outlet. Preliminary treatment and high-rate 

disinfection were selected as the most economical method to meet the treatment 

objectives for this project. 

Operational Requirements. The ORF are designed to operate on a continuous basis 

include: 

• Preliminary Treatment 

• High-Rate Disinfection 

Each of the five ORF tanks fill uniformly during an overflow event. After the wetweather flows subside, the flow retained in the ORF is pumped to the WWTF 

headworks for treatment. WWTF staff manually controls this operation. The tanks are 

flushed by a manually controlled system using chlorinated plant effluent. 

Chemical Storage and Feed Requirements. The ORF utilize two chemicals: 

• Liquid Sodium Hypochlorite for disinfection 

• Liquid Sodium Bisulfite for dechlorination 

The liquid sodium hypochlorite system design is based on commercial-grade liquid 

sodium hypochlorite (15% solution) which contains the equivalent of 160 kg of 

chlorine/m3

 (1.25 lbs Cl2/gal) of solution. The metering pumps are designed to deliver a 

dosage of 25 mg/L at the peak design flow with one unit out of service. The liquid 

sodium hypochlorite is stored in a bulk storage tank with an effective working volume of 

10 m3

 (2,500 gal). The available chlorine in sodium hypochlorite solution declines with 

age. The anticipated sodium hypochlorite storage time will be 51 days initially, 

decreasing to 25 days at the future design flow. 

The liquid sodium bisulfite storage system design is based on commercial grade liquid 

sodium bisulfite (38% solution) that contains appropriately 503 kg of sodium bisulfite/m3 

(4.2 lbs NaHSO3/gal) of solution. The metering pumps are designed to deliver sodium 

bisulfite dosages necessary to dechlorinate effluent with 1.5 mg/L of sodium bisulfite per 

mg/L of chlorine residual. Approximately 1.6 kg of sodium bisulfite will be used per kg 

of chlorine residual. A bulk storage tank with an effective working volume of 3.8 m3 

(1,000 gal) is provided. 

The sodium hypochlorite and sodium bisulfite metering pumps are flow paced. An 

ultrasonic type level sensor is used for effluent flow measurement. A broad-crested 

rectangular weir 3 m (10 ft) in length is provided as the primary measuring device. 

182
 
Sludge Handling Requirements. Accumulated sludge will be handled through the pump 

back of retained WWF and the associated tank flushing following tank draining. 

Operation and Maintenance Requirements. The implementation of the ORF will result 

in increased O&M costs. The following list summarizes the areas of expected O&M 

cost increases at the WWTF: 

• Chemicals for disinfection and dechlorination 

• Power for tank draining and flushing, and chemical feed systems. 

The estimated annual O&M costs are provided in Table 5.7.a-3. 

Table 5.7.a-3. ORF O&M Costs 

Annual Chemical Costs Cost 

Sodium Hypochlorite $7,000 

Sodium Bisulfite $3,000 

O&M $8,000 

Total $18,000 

Project Costs
 

Project costs for the capital costs for the ORF are provided in Table 5.7.a-4.
 

Table 5.7.a-4. ORF Chlorination/dechlorination Capital Costs
 

Chlorination Cost 

Chemical Feed Pumps and Controls $20,000 

Storage Facilities $14,000 

Installation $21,000 

Subtotal $55,000 

Dechlorination 

Chemical Feed Pumps and Controls $20,000 

Storage Facilities $16,000 

Residual Analyzer System $10,000 

Installation $19,000 

Subtotal $65,000 

Total Cost $120,000 

183
 
The conventional treatment alternative for this retrofit would be to construct wet-weather 

storage facilities to equalize WWF for ultimate treatment at the WWTF. This alternative 

would require the construction of a 4,100 m3

 (1.083 MG) ORF. The estimated cost to 

construct this facility based upon EPA cost guidelines for CSO control technologies 

(EPA 1992) is $3.816 million derived from the following equation: 

Cost = 3.577 V0.812 (1) 

where, 

Cost = Average construction cost, millions of dollars 

V = Storage volume, MG 

The ORF retrofit is a simple solution, requiring minimal O&M cost, resulting in a total 

cost that is a fraction of the cost of constructing new wet-weather storage facilities. The 

ORF retrofit provides an average annual 76% wet-weather volume capture. The cost for 

this retrofit is $120,000 dollars compared to the $3,816,000 cost for conventional 

storage facilities. 

Conclusions 

Retrofitting the abandoned primary treatment tanks for use as ORF was a success, in 

large part due to the availability and condition of the existing facilities. The abandoned 

primary tanks were in good condition and required only minor modifications for use as 

an ORF. The ORF retrofit is a cost-effective alternative in cases where abandoned 

primary tanks are available in good condition for retrofitting. This type of retrofit may be 

a less cost-effective solution for cases in where the existing tanks are in poor condition 

requiring a substantial cost to upgrade the tanks to serve a functional role as an ORF. 

184
 
Section 6 – References 

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APPENDIX A

BIBLIOGRAPHIC DATABASES

Wilson Applied Science & Technology Abstracts (WBA) 

H.W. Wilson Company Phone: (718) 588-8400

950 University Ave. 800-367-6770

Bronx, NY 10452 Fax (718) 538-2716

Contact: Technical Support Department.

E-mail: techmailinformationo.hwwiison.com (Internet).

Type: Bibliographic,

Content: Contains abstracts to all articles cited in the Applied Science Technology

Index (described in a separate entry) from March 1993 to date. Includes an index of

non-English-language periodicals English abstracts are provided. Sources include 400

English language scientific and technical publications.

Subject Coverage. Applied science and technology, including such areas as

aeronautics and space science, chemistry, computer technology, construction, data

processing, energy, engineering, the environment, fire prevention, food, geology,

machinery, marine technology and oceanography, mathematical metallurgy,

meteorology, mineralogy, optics, petroleum and gas, physics, robotics, solid state

technology, telecommunications, textiles, transportation, and water and waste.

Language: English. 

Geographic Coverage: International. 

Time Span: October 1983 to date (indexing coverage); March 1993 to date (Abstracts). 

Updating: Approximately 1,000 records per week. 

191
Compendex*Plus

Engineering Information, Inc. (Ei) Phone: (201) 216-8500,

Castle Point on the Hudson 800-221-1044

Hoboken, NJ 07030 Fax: (201) 216-8532

Type: Bibliographic.

Content: Contains more than 3.63 million citations, with abstracts, from more than

2,500 journals, reports, books, and conference proceedings covering fields of

engineering and technology. Sources include approximately 2500 journals, reports,

books, and conference proceedings. Conference review records contain conference

code numbers that are links to complete coverage of all individual papers. Corresponds

in part to The Engineering Index Monthly and Engineering Index Annual.

Subject Coverage: All fields of engineering, including related areas of app science,

management, and energy. Specific fields covered include civil engineering,

construction and materials, transportation, waterworks, sanitary, engineering,

bioengineering, ocean engineering, mining, petroleum, fuel technology, metallurgical

engineering, mechanical engineering, aerospace, engineering, automotive engineering,

electronics and electrical engineering, control engineering, chemical engineering,

nuclear engineering, and computers and robotics engineering.

Language: English. 

Geographic Coverage: International. 

Time Span: 1970 to date. 

Updating: Weekly; 13,000 records a month. 

192
Environmental Sciences & Pollution Management Database 

Cambridge Scientific Abstracts (CSA) Phone: (301)961-6750 

7200 Wisconsin Ave., Ste. 601 800-843-7751 

Bethesda, MD 20814-4823 Fax: (301)961-6720 

Type: Bibliographic. 

Content: Contains more than 64,000 citations, with abstracts, to literature covering the 

environmental sciences. Covers some 1498 core journals, monographs, and 

conference proceedings, as well as 5500 secondary sources, drawn from the primary 

sources of abstracts journals such as: Ecology, Pollution Digest of Environmental 

Impact Statements, toxicology, Health & Safety Science, Bacteriology, Environmental 

Engineering, and others. 

Subject Coverage: Environmental sciences, including agricultural biotechnology, 

aquatic environments, bacteriology, ecology, environmental biotechnology, 

environmental engineering, environmental impact statements, industrial hygiene, 

microbiology applied to industrial and environmental issues, pollution (land, air, water, 

noise, and solid waste), risk assessment, safety science, toxicology, water quality, and 

water resource issues. 

Language: English. 

Geographic Coverage: United States. 

Time Span: 1983 to date. 

Updating: Monthly. 76,DOO records/year. 

National Center for Environmental Publications and Information 

United States Environmental Protection Agency 

Washington, D.C. 

http://www.epa.gov/ncepihom/index.html 

Type: National Publications Catalog 

Content: 5,500 titles in paper and/or electronic. 

Subject Coverage: Repository for all EPA documents. 

Language: English. 

Time Span: 

Updating: Yearly. 

193
ProCite 

Robert E. Pitt, P.E., Ph.D., DEE

University of Alabama at Birmingham

Hoehn Engineering Building

1075 - 13th St. So.

Birmingham. AL 35294

Type: Bibliographic.

Phone: (205) 934-8434

Fax: (205) 975-9042

e-mail: RPITT@ENG.UAB.EDU

Content: 3,700 wet-weather references. 

Subject Coverage: The database was prepared by Robert Pitt for the Wet Weather 

Floe Design for the Future project under contract by the United States Environmental 

Protection Agency. 

Language: English. 

Time Span: 1899-1997 

Updating: Database constructed in 1997. 

194
APPENDIX B

SELECTED TABLES WITH ENGLISH EQUIVALENT UNITS

195
Table 5.1.b-2. Comparison of ACTIFLO and Existing Primary Treatment Systems 

Net Savings with 

Parameter Existing System ACTIFLO ACTIFLO 

Total Capacity 240 MG

Number of Units 8

Unit Capacity 30 MGD

Area Required 14,307 ft2

 (per unit)

114,453 ft2

 (total) 

Process Volume 143.066 ft3

 (per unit) 

1,114,530 ft3

 (total) 

Hydraulic Retention 51 min 

Time 

240 MG ­

6 ­

40 MGD ­

945 ft2

 (per train) 94% per unit 

5,670 ft2

 (total) 95% total 

20,790 ft3

 (per unit) 85% per unit 

124,740 ft3

 (total) 89% per total 

5.6 min 89% 

196