DISCLAIMER 

This document provides recommended criteria for the postearthquake damage assessment, 

evaluation and repair of steel moment-frame buildings. These recommendations were developed 

by practicing engineers, based on professional judgment and experience, and by a program of 

laboratory, field and analytical research. It is primarily intended as a resource for communities in 

developing formal postearthquake damage assessment and repair programs, but may also be used as 

a resource by individual engineers and building officials, when such formal programs have not been 

adopted. While every effort has been made to solicit comments from a broad selection of the affected 

parties, this is not a consensus document. No warranty is offered, with regard to the 

recommendations contained herein, either by the Federal Emergency Management Agency, 

the SAC Joint Venture, the individual Joint Venture partners, or their directors, members or 

employees. These organizations and their employees do not assume any legal liability or 

responsibility for the accuracy, completeness, or usefulness of any of the information, 

products or processes included in this publication. The reader is cautioned to review 

carefully the material presented herein and exercise independent judgment as to its 

suitability for application to specific engineering projects. These recommended criteria have 

been prepared by the SAC Joint Venture with funding provided by the Federal Emergency 

Management Agency, under contract number EMW-95-C-4770. 

Cover Art. The beam-column connection assembly shown on the cover depicts the standard 

detailing used in welded steel moment-frame construction prior to the 1994 Northridge 

earthquake. This connection detail was routinely specified by designers in the period 1970-1994 

and was prescribed by the Uniform Building Code for seismic applications during the period 

1985-1994. It is no longer considered to be an acceptable design for seismic applications. 

Following the Northridge earthquake, it was discovered that many of these beam-column 

connections had experienced brittle fractures at the joints between the beam flanges and column 

flanges. 
Recommended Postearthquake Evaluation and Repair 

Criteria for Welded Steel Moment-Frame Buildings 

SAC Joint Venture 

A partnership of

Structural Engineers Association of California (SEAOC)

Applied Technology Council (ATC)

California Universities for Research in Earthquake Engineering (CUREe)

Prepared for SAC Joint Venture Partnership by

Guidelines Development Committee

Ronald O. Hamburger, Chair 

John D. Hooper Thomas Sabol 

Robert Shaw C. Mark Saunders 

Lawrence D. Reaveley Raymond H. R. Tide 

Project Oversight Committee 

Shirin Ader

John M. Barsom

Roger Ferch

Theodore V. Galambos

John Gross

James R. Harris

Richard Holguin

William J. Hall, Chair 

Nestor Iwankiw 

Roy G. Johnston 

Leonard Joseph 

Duane K. Miller 

John Theiss 

John H. Wiggins 

SAC Project Management Committee 

SEAOC: William T. Holmes Program Manager: Stephen A. Mahin 

ATC: Christopher Rojahn 

CUREe: Robin Shepherd 

Project Director for Topical Investigations: 

James O. Malley 

Project Director for Product Development: 

Ronald O. Hamburger 

SAC Joint Venture 

SEAOC: www.seaoc.org 

ATC: www.atcouncil.org 

CUREe: www.curee.org 

June, 2000 
THE SAC JOINT VENTURE 

SAC is a joint venture of the Structural Engineers Association of California (SEAOC), the Applied 

Technology Council (ATC), and California Universities for Research in Earthquake Engineering 

(CUREe), formed specifically to address both immediate and long-term needs related to solving 

performance problems with welded, steel moment-frame connections discovered following the 1994 

Northridge earthquake. SEAOC is a professional organization composed of more than 3,000 practicing 

structural engineers in California. The volunteer efforts of SEAOC’s members on various technical 

committees have been instrumental in the development of the earthquake design provisions contained in 

the Uniform Building Code and the 1997 National Earthquake Hazards Reduction Program (NEHRP) 

Recommended Provisions for Seismic Regulations for New Buildings and Other Structures. ATC is a 

nonprofit corporation founded to develop structural engineering resources and applications to mitigate 

the effects of natural and other hazards on the built environment. Since its inception in the early 1970s, 

ATC has developed the technical basis for the current model national seismic design codes for buildings; 

the de-facto national standard for postearthquake safety evaluation of buildings; nationally applicable 

guidelines and procedures for the identification, evaluation, and rehabilitation of seismically hazardous 

buildings; and other widely used procedures and data to improve structural engineering practice. CUREe 

is a nonprofit organization formed to promote and conduct research and educational activities related to 

earthquake hazard mitigation. CUREe’s eight institutional members are the California Institute of 

Technology, Stanford University, the University of California at Berkeley, the University of California at 

Davis, the University of California at Irvine, the University of California at Los Angeles, the University 

of California at San Diego, and the University of Southern California. These university earthquake 

research laboratory, library, computer and faculty resources are among the most extensive in the United 

States. The SAC Joint Venture allows these three organizations to combine their extensive and unique 

resources, augmented by consultants and subcontractor universities and organizations from across the 

nation, into an integrated team of practitioners and researchers, uniquely qualified to solve problems 

related to the seismic performance of steel moment-frame structures. 

ACKNOWLEDGEMENTS 

Funding for Phases I and II of the SAC Steel Program to Reduce the Earthquake Hazards of Steel 

Moment-Frame Structures was principally provided by the Federal Emergency Management Agency, 

with ten percent of the Phase I program funded by the State of California, Office of Emergency Services. 

Substantial additional support, in the form of donated materials, services, and data has been provided by 

a number of individual consulting engineers, inspectors, researchers, fabricators, materials suppliers and 

industry groups. Special efforts have been made to maintain a liaison with the engineering profession, 

researchers, the steel industry, fabricators, code-writing organizations and model code groups, building 

officials, insurance and risk-management groups, and federal and state agencies active in earthquake 

hazard mitigation efforts. SAC wishes to acknowledge the support and participation of each of the above 

groups, organizations and individuals. In particular, we wish to acknowledge the contributions provided 

by the American Institute of Steel Construction, the Lincoln Electric Company, the National Institute of 

Standards and Technology, the National Science Foundation, and the Structural Shape Producers 

Council. SAC also takes this opportunity to acknowledge the efforts of the project participants – the 

managers, investigators, writers, and editorial and production staff – whose work has contributed to the 

development of these documents. Finally, SAC extends special acknowledgement to Mr. Michael 

Mahoney, FEMA Project Officer, and Dr. Robert Hanson, FEMA Technical Advisor, for their continued 

support and contribution to the success of this effort. 
Recommended Postearthquake Evaluation 

and Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings Table of Contents 

TABLE OF CONTENTS 

LIST OF FIGURES ..............................................................................................................................ix

LIST OF TABLES................................................................................................................................xi

1. INTRODUCTION............................................................................................................ 1-1

1.1 Purpose................................................................................................................. 1-1

1.2 Intent .................................................................................................................... 1-2

1.3 Background .......................................................................................................... 1-4

1.4 Application ........................................................................................................ 1-12

1.5 Postearthquake Evaluation and Repair Process ................................................. 1-12

1.6 Overview of These Recommended Criteria 1-14

2. INSPECTION AND CLASSIFICATION OF DAMAGE ............................................... 2-1

2.1 Introduction.......................................................................................................... 2-1

2.2 Damage Types...................................................................................................... 2-1

2.2.1 Girder Damage......................................................................................... 2-2

2.2.2 Column Flange Damage........................................................................... 2-4

2.2.3 Weld Damage........................................................................................... 2-6

2.2.4 Shear Tab Damage ................................................................................... 2-8

2.2.5 Panel Zone Damage ................................................................................. 2-9

2.2.6 Other Damage ........................................................................................ 2-10

3. PRELIMINARY POSTEARTHQUAKE ASSSESSMENT............................................ 3-1

3.1 Introduction.......................................................................................................... 3-1

3.1.1 General..................................................................................................... 3-1

3.1.2 Evaluator Qualifications .......................................................................... 3-2

3.2. Screening.............................................................................................................. 3-2

3.3 Preliminary Evaluation ........................................................................................ 3-4

3.3.1 General..................................................................................................... 3-4

3.3.2 Building Construction Characteristics ..................................................... 3-6

3.3.3 Preliminary Site Inspection...................................................................... 3-6

3.3.3.1 Preliminary Connection Inspections when

Fireproofing is Present.............................................................. 3-9

3.3.3.2 Bare Structural Steel ............................................................... 3-10

3.3.4 Data Reduction and Assessment............................................................ 3-12

3.3.4.1 Finding of Dangerous Condition............................................. 3-12

3.3.4.2 Finding of Damaged Condition.................................................. 3-12

3.3.4.3 Finding of Undamaged Condition........................................... 3-13

3.3.5 Reporting and Notification .................................................................... 3-13

3.3.5.1 Building Departments ............................................................. 3-14

3.3.5.2 Private Consultants ................................................................. 3-14

4. LEVEL 1 DETAILED POSTEARTHQUAKE EVALUATIONS .................................. 4-1

4.1 Introduction.......................................................................................................... 4-1

iii 
Recommended Postearthquake Evaluation 

FEMA-352 and Repair Criteria for Welded 

Table of Contents Steel Moment-Frame Buildings 

4.2 Data Collection .................................................................................................... 4-2

4.3 Evaluation Approach 4-2

4.4 Detailed Procedure............................................................................................... 4-3

4.4.1 Method 1 – Inspection of All Connections .............................................. 4-4

4.4.1.1 Detailed Connection Inspections 4-4

4.4.1.1.1 Initial Inspections 4-6

4.4.1.1.2 Detailed Inspections................................................... 4-7

4.4.1.2 Damage Characterization.......................................................... 4-7

4.4.1.3 Determine Damage Index at Each Floor

for Each Direction of Response .............................................. 4-10

4.4.1.4 Determine Maximum Floor Damage Index ............................ 4-11

4.4.1.5 Determine Recommended Recovery Strategies

for the Building....................................................................... 4-11

4.4.2 Method 2 – Inspection of a Sample of Connections ........................... 4-12

4.4.2.1 Evaluation Step 1 – Categorize Connections by Groups ........ 4-13

4.4.2.2 Step 2 – Select Samples of Connections for Inspection.......... 4-13

4.4.2.2.1 Method A – Random Selection................................ 4-14

4.4.2.2.2 Method B – Analytical Selection ............................. 4-17

4.4.2.3 Step 3 – Inspect the Selected Samples of Connections........... 4-18

4.4.2.3.1 Inspection................................................................. 4-18

4.4.2.3.2 Damage Characterization......................................... 4-19

4.4.2.4 Step 4 – Inspect Connections Adjacent to Damaged 

Connections............................................................................. 4-20

4.4.2.5 Step 5 – Determine Damage Statistics for Each Group .......... 4-21

4.4.2.6 Step 6 – Determine the Probability that the Connections

in a Group at a Floor Level Sustained Excessive Damage ..... 4-21

4.4.2.7 Step 7 – Determine Recommended Recovery Strategies

for the Building....................................................................... 4-24

4.4.3 Additional Considerations ..................................................................... 4-25

4.5. Evaluation Report .............................................................................................. 4-25

5. LEVEL 2 DETAILED POSTEARTHQUAKE EVALUATIONS .................................. 5-1

5.1 Introduction.......................................................................................................... 5-1

5.2 Data Collection .................................................................................................... 5-2

5.3 Evaluation Approach ........................................................................................... 5-3

5.4 Field Inspection.................................................................................................... 5-6

5.5 Material Properties and Condition Assessment ................................................... 5-6

5.5.1 Material Properties................................................................................... 5-7

5.6 Structural Performance Confidence Evaluation................................................. 5-12

5.7 Ground Motion Representation.......................................................................... 5-15

5.7.1 Instrumental Recordings ........................................................................ 5-15

5.7.2 Estimated Ground Motion...................................................................... 5-16

5.8 Analytical Procedures ........................................................................................ 5-17

5.8.1 Procedure Selection ............................................................................... 5-18

5.8.2 Linear Static Procedure (LSP) ............................................................... 5-18

iv 
Recommended Postearthquake Evaluation 

and Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings Table of Contents 

5.8.2.1 Basis of the Procedure............................................................. 5-18

5.8.2.2 Modeling and Analysis Considerations .................................. 5-21

5.8.2.2.1 Period Determination............................................... 5-21

5.8.2.3 Determination of Actions and Deformations.......................... 5-21

5.8.2.3.1 Pseudo Lateral Load................................................. 5-21

5.8.2.3.2 Vertical Distribution of Seismic Forces................... 5-23

5.8.2.3.3 Horizontal Distribution of Seismic Forces .............. 5-23

5.8.2.3.4 Determination of Interstory Drift ............................. 5-23

5.8.2.3.5 Determination of Column Demands ........................ 5-24

5.8.3 Linear Dynamic Procedure (LDP) ......................................................... 5-25

5.8.3.1 Basis of the Procedure............................................................. 5-25

5.8.3.2 Modeling and Analysis Considerations .................................. 5-26

5.8.3.2.1 General..................................................................... 5-26

5.8.3.3 Determination of Actions and Deformations.......................... 5-26

5.8.3.3.1 Factored Interstory Drift Demand............................ 5-26

5.8.3.3.2 Determination of Column Demands ........................ 5-26

5.8.4 Nonlinear Static Procedure (NSP) ......................................................... 5-26

5.8.4.1 Basis of the Procedure............................................................. 5-26

5.8.4.2 Modeling and Analysis Considerations .................................. 5-27

5.8.4.2.1 General..................................................................... 5-27

5.8.4.2.2 Control Node............................................................ 5-28

5.8.4.2.3 Lateral Load Patterns ............................................... 5-28

5.8.4.2.4 Period Determination............................................... 5-28

5.8.4.2.5 Analysis of Three-Dimensional Models .................. 5-28

5.8.4.2.6 Analysis of Two-Dimensional Models .................... 5-28

5.8.4.3 Determination of Actions and Deformations ......................... 5-29

5.8.4.3.1 Target Displacement ................................................ 5-29

5.8.4.3.2 Diaphragms .............................................................. 5-29

5.8.4.3.3 Factored Interstory Drift Demand............................ 5-29

5.8.4.3.4 Factored Column and Column Splice Demands...... 5-29

5.8.5 Nonlinear Dynamic Procedure (NDP) ................................................... 5-29

5.8.5.1 Basis of the Procedure............................................................. 5-29

5.8.5.2 Modeling and Analysis Assumptions ..................................... 5-30

5.8.5.2.1 General .................................................................... 5-30

5.8.5.2.2 Ground Motion Characterization ............................. 5-30

5.8.5.3 Determination of Actions and Deformations.......................... 5-30

5.8.5.3.1 Response Quantities................................................. 5-30

5.8.5.3.2 Factored Interstory Drift Demand............................ 5-30

5.8.5.3.3 Factored Column and Column Splice Demands...... 5-30

5.9 Mathematical Modeling..................................................................................... 5-31

5.9.1 Modeling Approach ............................................................................... 5-31

5.9.2 Model Configuration.............................................................................. 5-31

5.9.3 Horizontal Torsion................................................................................. 5-32

5.9.4 Foundation Modeling............................................................................. 5-32

v 
Recommended Postearthquake Evaluation 

FEMA-352 and Repair Criteria for Welded 

Table of Contents Steel Moment-Frame Buildings 

5.9.5 Diaphragms ............................................................................................ 5-33

5.9.6 P-D effects.............................................................................................. 5-33

5.9.7 Elastic Framing Properties..................................................................... 5-35

5.9.8 Nonlinear Framing Properties................................................................ 5-35

5.9.9 Verification of Analysis Assumptions ................................................... 5-36

5.9.10 Undamaged Connection Modeling ....................................................... 5-36

5.9.10.1 Fully Restrained Connections ................................................. 5-36

5.9.10.1.1 Linear Modeling....................................................... 5-36

5.9.10.1.2 Nonlinear Modeling................................................. 5-37

5.9.10.2 Simple Shear Tab Connections............................................... 5-37

5.9.10.2.1 Modeling Guidelines – Linear Analysis .................. 5-38

5.9.10.2.2 Modeling Guidelines – Nonlinear Analysis............. 5-38

5.9.11 Damage Modeling...................................................................... 5-38

5.9.11.1 Fully Restrained (FR) Connection Damage............................ 5-39

5.9.11.2 Column Damage ..................................................................... 5-39

5.9.11.3 Beam Damage ......................................................................... 5-40

5.9.11.4 Other Damage ......................................................................... 5-40

5.10 Acceptance Criteria and Confidence Evaluation ............................................... 5-40

5.10.1 Factored Demand to Capacity Ratio ...................................................... 5-41

5.10.2 Performance Limited by Interstory Drift Angle..................................... 5-43

5.10.2.1 Factored Interstory Drift Angle Demand ............................. 5-43

5.10.2.2 Factored Interstory Drift Angle Capacity............................. 5-45

5.10.2.2.1 Global Interstory Drift Angle................................... 5-45

5.10.2.2.2 Local Interstory Drift Angle..................................... 5-46

5.10.3 Performance Limited by Column Compressive Capacity...................... 5-47

5.10.3.1 Column Compressive Demand ............................................ 5-47

5.10.3.2 Column Compressive Capacity............................................ 5-48

5.10.4 Column Splice Capacity......................................................................... 5-49

5.10.4.1 Column Splice Tensile Demand .......................................... 5-49

5.10.4.2 Column Splice Tensile Capacity.......................................... 5-49

5.11 Evaluation Report .............................................................................................. 5-49

6. POSTEARTHQUAKE REPAIR ..................................................................................... 6-1

6.1 Scope 6-1

6.2 Shoring and Temporary Bracing.......................................................................... 6-2

6.2.1 Investigation............................................................................................. 6-2

6.2.2 Special Requirements............................................................................... 6-2

6.3 Repair Details....................................................................................................... 6-3

6.3.1 Approach.................................................................................................. 6-3

6.3.2 Weld Fractures – Type W Damage.......................................................... 6-4

6.3.3 Column Fractures – Types C1 to C5 and P1 to P6 .................................. 6-7

6.3.4 Column Splice Fractures – Type C7 ...................................................... 6-11

6.3.5 Girder Flange Fractures – Types G3 to G5............................................ 6-11

6.3.6 Buckled Girder Flanges – Type G1........................................................ 6-13

vi 
Recommended Postearthquake Evaluation 

and Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings Table of Contents 

6.3.7 Buckled Column Flanges – Type C6 ..................................................... 6-13

6.3.8 Gravity Connections .............................................................................. 6-14

6.3.9 Reuse of Bolts........................................................................................ 6-14

6.3.10 Welding Specifications .......................................................................... 6-15

6.4 Preparation ......................................................................................................... 6-15

6.4.1 Welding Procedure Specifications......................................................... 6-15

6.4.2 Welder Training..................................................................................... 6-16

6.4.3 Welder Qualifications ............................................................................ 6-16

6.4.4 Joint Mock-Ups...................................................................................... 6-16

6.4.5 Repair Sequence..................................................................................... 6-17

6.4.6 Concurrent Work ................................................................................... 6-17

6.5 Execution ........................................................................................................... 6-17

6.5.1 General................................................................................................... 6-17

6.5.2 Removal of Backing............................................................................... 6-18

6.5.3 Removal of Weld Tabs .......................................................................... 6-19

6.5.4 Defect Removal ..................................................................................... 6-19

6.5.5 Girder Repair ......................................................................................... 6-20

6.5.6 Weld Repair (Types W2, or W3 and Defects) ....................................... 6-21

6.5.7 Weld Overlays........................................................................................ 6-21

6.5.8 Column Flange Repairs – Type C2........................................................ 6-23

APPENDIX A. DETAILED PROCEDURES FOR PERFORMANCE EVALUATION......... A-1

A.1 Scope .................................................................................................................. A-1

A.2 Performance Evaluation Approach ..................................................................... A-1

A.2.1 Confidence of Ability to Withstand Collapse......................................... A-1

A.2.2 Basic Procedure....................................................................................... A-3

A.3 Determination of Hazard Parameters.................................................................. A-6

A.4 Determination of Demand Factors...................................................................... A-6

A.5 Determination of Beam-Column Connection Assembly Capacities................... A-9

A.5.1 Connection Test Protocols .................................................................... A-10

A.5.2 Determination of Beam-Column Assembly Capacities

and Resistance Factors.......................................................................... A-10

A.6 Global Stability Capacity.................................................................................. A-11

APPENDIX B. SAMPLE PLACARDS.....................................................................................B-1

B.1 “Inspected” Placard..............................................................................................B-1

B.2 “Restricted Use” Placard......................................................................................B-2

B.3 Modified “Restricted Use” Placard......................................................................B-3

B.4 “Unsafe” Placard..................................................................................................B-4

APPENDIX C. SAMPLE INSPECTION FORMS ....................................................................C-1

REFERENCES, BIBLIOGRAPHY, AND ACRONYMS...........................................................R-1

SAC PROJECT PARTICIPANTS...............................................................................................S-1

vii 

Recommended Postearthquake Evaluation 

and Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings List of Figures 

LIST OF FIGURES

Figure 1-1 Typical Welded Moment-Resisting Connection Prior to 1994............................ 1-6

Figure 1-2 Common Zone of Fracture Initiation in Beam-Column Connection ................... 1-7

Figure 1-3 Fractures of Beam to Column Joints.................................................................... 1-7

Figure 1-4 Column Fractures................................................................................................. 1-7

Figure 1-5 Vertical Fracture through Beam Shear Plate Connection .................................... 1-8

Figure 1-6 Flow Chart for Postearthquake Actions............................................................. 1-13

Figure 2-1 Elements of Welded Steel Moment Frame .......................................................... 2-1

Figure 2-2 Types of Girder Damage...................................................................................... 2-3

Figure 2-3 Types of Column Damage ................................................................................... 2-4

Figure 2-4 Types of Weld Damage........................................................................................ 2-6

Figure 2-5 Types of Shear Tab Damage ................................................................................ 2-8

Figure 2-6 Types of Panel Zone Damage .............................................................................. 2-9

Figure 3-1 Observation Zones for Fire-Proofed Beam-Column Connections....................... 3-9

Figure 3-2 Components of Moment Connection................................................................. 3-11

Figure 4-1 Fireproofing Removal for Initial Connection Inspection..................................... 4-5

Figure 4-2 Fireproofing Removal for Complete Connection Inspection............................... 4-5

Figure 4-3 Inspection of Connections Adjacent to Damaged Connection (1 £ dj £ 2)........ 4-20

Figure 4-4 Inspection of Connections Adjacent to Damaged Connection (dj ‡ 3).............. 4-20

Figure 5-1 Presumed Postearthquake Hazard Curve ........................................................... 5-14

Figure 5-2 Welded Unreinforced Fully Restrained Connection (pre-1994)........................ 5-36

Figure 5-3 Typical Simple Shear Tab Connection with Slab .............................................. 5-37

Figure 5-4 Type P9 Panel Zone Damage............................................................................. 5-40

Figure 6-1 Gouge and Re-weld of Root Defect or Damage .................................................. 6-5

Figure 6-2 Gouge and Re-weld of Fractured Weld ............................................................... 6-5

Figure 6-3 Backgouge and Reweld Repair ............................................................................ 6-6

Figure 6-4 Temporary Removal of Beam Section for Access............................................... 6-8

Figure 6-5 Backgouge and Reweld of Column Flange.......................................................... 6-9

Figure 6-6 Replacement of Column Flange Repair............................................................... 6-9

Figure 6-7 Reweld Repair of Web Plate and Doubler Plate................................................ 6-10

Figure 6-8 Alternative Column Web Repair – Columns without Doubler Plates ............... 6-10

Figure 6-9 Beam Flange Plate Replacement........................................................................ 6-12

Figure 6-10 Alternative Beam Flange Plate Replacement..................................................... 6-12

Figure 6-11 Addition of Stiffeners at Buckled Girder Flange ............................................... 6-13

Figure 6-12 Weld Overlay Repair of Beam Flange to Column Flange Joint ........................ 6-22

Figure 6-13 Plan View and Assumed Stress Distribution for Weld Overlay Design............ 6-23

Figure A-1 Representative Incremental Dynamic Analysis Plots ....................................... A-12

Figure C-1 Inspection Form – Major Axis Column Connection............................................C-1

Figure C-2 Inspection Form – Large Discontinuities – Major Axis.......................................C-2

Figure C-3 Inspection Form – Minor Axis Column Connection ...........................................C-3

Figure C-4 Inspection Form – Large Discontinuities – Minor Axis ......................................C-4

ix 

Recommended Postearthquake Evaluation 

and Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings List of Tables 

LIST OF TABLES

Table 2-1 Types of Girder Damage...................................................................................... 2-2

Table 2-2 Types of Column Damage ................................................................................... 2-4

Table 2-3 Types of Weld Damage, Defects and Discontinuities ......................................... 2-6

Table 2-4 Types of Shear Tab Damage ................................................................................ 2-8

Table 2-5 Types of Panel Zone Damage .............................................................................. 2-9

Table 3-1 Ground Motion Indicators of Potential Damage.................................................. 3-3

Table 3-2 Postearthquake Condition Designations .............................................................. 3-5

Table 4-1a Connection Damage Indices ................................................................................ 4-9

Table 4-1b Connection Damage Indices for Common Damage Combinations ................... 4-10

Table 4-2 Recommended Repair and Modification Strategies........................................... 4-11

Table 4-3 Minimum Sample Size for Connection Groups................................................. 4-15

Table 4-4 Pf as a Function of Parameter b.......................................................................... 4-23

Table 4-5 Recommended Condition Designation and Repair Strategies ........................... 4-24

Table 5-1 Default Material Specifications for Steel Moment-Frame Buildings .................. 5-8

Table 5-2 Lower Bound and Expected Material Properties for Structural Steel Shapes

of Various Grades ................................................................................................ 5-9

Table 5-3 Recommended Occupancy Actions, Based on Detailed Evaluation.................. 5-13

Table 5-4 Selection Criteria for Analysis Procedure to Achieve Collapse Prevention ...... 5-19

Table 5-5 Modification Factors C3 for Linear Static Procedure......................................... 5-23

Table 5-6 Performance Parameters Requiring Evaluation of Confidence ......................... 5-41

Table 5-7 Factored-Demand-to-Capacity Ratios l and Uncertainty bUT, for

Specific Confidence Levels ............................................................................... 5-42

Table 5-8 Interstory Drift Angle Analysis Demand Uncertainty Factors, ga ...................... 5-44

Table 5-9 Interstory Drift Angle Analysis Demand Variability Factors, g,

Type 1 and Type 2 Connections......................................................................... 5-44

Table 5-10 Global Interstory Drift Angle Capacity and Resistance Factors ........................ 5-45

Table 5-11 Uncertainty Coefficient bUT for Global Interstory Drift Evaluation .................. 5-46

Table 5-12 Local Interstory Drift Angle Capacity and Resistance Factors .......................... 5-46

Table 5-13 Uncertainty Coefficient bUT for Local Interstory Drift Evaluation .................... 5-47

Table 5-14 Analysis Uncertainty Factor ga and Total Uncertainty Coefficient bUT for

Evaluation of Column Compressive Demands.................................................. 5-48

Table 6-1 Reference Details for Type W Damage ............................................................... 6-4

Table 6-2 Reference Details for Type C and P Damage ...................................................... 6-8

Table A-1 Default Logarithmic Uncertainty bDU for Various Analytical Methods.............. A-8

Table A-2 Default Bias Factors CB ...................................................................................... A-8

xi 
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings Chapter 1: Introduction 

1. INTRODUCTION 

1.1 Purpose 

This report, FEMA-352 – Recommended Postearthquake Evaluation and Repair Criteria for 

Welded Steel Moment-Frame Buildings, has been developed by the SAC Joint Venture under 

contract to the Federal Emergency Management Agency (FEMA) to provide communities and 

organizations developing programs for the assessment, occupancy status, and repair of welded 

steel moment-frame buildings that have been subjected to the effects of strong earthquake ground 

shaking. It is one of a series of companion publications addressing the issue of the seismic 

performance of steel moment-frame buildings. The set of companion publications includes: 

• FEMA-350 – Recommended Seismic Design Criteria for New Steel Moment-Frame 

Buildings. This publication provides recommended criteria, supplemental to FEMA 302 – 

1997 NEHRP Recommended Provisions for Seismic Regulations for New Buildings and other 

Structures, for the design and construction of steel moment-frame buildings and provides 

alternative performance-based design criteria. 

• FEMA-351 – Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded 

Steel Moment-Frame Buildings. This publication provides recommended methods to 

evaluate the probable performance of existing steel moment-frame buildings in future 

earthquakes and to retrofit these buildings for improved performance. 

• FEMA-352 – Recommended Postearthquake Evaluation and Repair Criteria for Welded Steel 

Moment-Frame Buildings. This publication provides recommendations for performing 

postearthquake inspections to detect damage in steel moment-frame buildings following an 

earthquake, evaluating the damaged buildings to determine their safety in the postearthquake 

environment, and repairing damaged buildings. 

• FEMA-353 – Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications. This publication provides 

recommended specifications for the fabrication and erection of steel moment frames for 

seismic applications. The recommended design criteria contained in the other companion 

documents are based on the material and workmanship standards contained in this document, 

which also includes discussion of the basis for the quality control and quality assurance 

criteria contained in the recommended specifications. 

The information contained in these recommended postearthquake damage assessment and 

repair criteria, hereinafter referred to as Recommended Criteria, is presented in the form of specific 

damage assessment, safety evaluation and repair procedures together with supporting commentary 

explaining part of the basis for these recommendations. Detailed derivations and explanations of 

the basis for these engineering recommendations may be found in a series of State of the Art 

Reports prepared in parallel with these Recommended Criteria. These reports include: 

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Chapter 1: Introduction Steel Moment-Frame Buildings 

• FEMA-355A – State of the Art Report on Base Metals and Fracture. This report summarizes 

current knowledge of the properties of structural steels commonly employed in building 

construction, and the production and service factors that affect these properties. 

• FEMA-355B – State of the Art Report on Welding and Inspection. This report summarizes 

current knowledge of the properties of structural welding commonly employed in building 

construction, the effect of various welding parameters on these properties, and the 

effectiveness of various inspection methodologies in characterizing the quality of welded 

construction. 

• FEMA-355C – State of the Art Report on Systems Performance of Steel Moment Frames 

Subject to Earthquake Ground Shaking. This report summarizes an extensive series of 

analytical investigations into the demands induced in steel moment-frame buildings designed 

to various criteria, when subjected to a range of different ground motions. The behavior of 

frames constructed with fully restrained, partially restrained and fracture-vulnerable 

connections is explored for a series of ground motions, including motion anticipated at nearfault and soft-soil sites. 

• FEMA-355D – State of the Art Report on Connection Performance. This report summarizes 

the current state of knowledge of the performance of different types of moment-resisting 

connections under large inelastic deformation demands. It includes information on fully 

restrained, partially restrained, and partial strength connections, both welded and bolted, 

based on laboratory and analytical investigations. 

• FEMA-355E – State of the Art Report on Past Performance of Steel Moment-Frame 

Buildings in Earthquakes. This report summarizes investigations of the performance of steel 

moment-frame buildings in past earthquakes, including the 1995 Kobe, 1994 Northridge, 

1992 Landers, 1992 Big Bear, 1989 Loma Prieta and 1971 San Fernando events. 

• FEMA-355F – State of the Art Report on Performance Prediction and Evaluation of Steel 

Moment-Frame Buildings. This report describes the results of investigations into the ability 

of various analytical techniques, commonly used in design, to predict the performance of 

steel moment-frame buildings subjected to earthquake ground motion. Also presented is the 

basis for performance-based evaluation procedures contained in the design criteria 

documents, FEMA-350, FEMA-351, and FEMA-352. 

In addition to the recommended design criteria and the State of the Art Reports, a companion 

document has been prepared for building owners, local community officials and other nontechnical audiences who need to understand this issue. A Policy Guide to Steel Moment-Frame 

Construction (FEMA 354), addresses the social, economic, and political issues related to the 

earthquake performance of steel moment-frame buildings. FEMA 354 also includes discussion 

of the relative costs and benefits of implementing the recommended criteria. 

1.2 Intent 

These Recommended Criteria are primarily intended as a resource document for communities 

developing formal programs for the assessment, occupancy status, and repair of buildings that have 

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Steel Moment-Frame Buildings Chapter 1: Introduction 

been subjected to the effects of strong earthquake ground shaking. They are also intended for direct 

use by engineers and building officials in communities without such formal programs. These 

criteria have been developed by professional engineers and researchers, based on the findings of 

a large multi-year program of investigation and research into the performance of steel momentframe buildings. Development of these recommended criteria was not subjected to a formal 

consensus review and approval process, nor was formal review or approval obtained from 

SEAOC’s technical committees. However, it did include broad external review by practicing 

engineers, researchers, fabricators, erectors, inspectors, building officials, and the producers of 

steel and welding consumables. In addition, two workshops were convened to obtain direct 

comment from these stakeholders on the proposed recommendations. 

The fundamental goal of the information presented in these Recommended Criteria is to assist 

the technical community in implementing effective programs for: 

• evaluation of steel moment-frame buildings affected by strong earthquake ground shaking to 

determine if they have been damaged, and to what extent, 

• identification of those buildings that have been so severely damaged that they constitute a 

significant safety hazard, and 

• repair of damaged structures such that they may safely be restored to long term occupancy. 

Commentary: When a severe earthquake effects a community, many buildings are 

likely to become damaged and some, as a result of this damage, may pose a 

significant safety hazard. In the past, building officials in such communities, in 

fulfillment of their charge to protect the public safety through regulation of 

building occupancy, have instituted programs of building inspection and posting 

to provide guidance to the public on the condition of affected structures and 

whether they should be entered. Depending on the individual community and its 

resources, the task of inspection and posting may be conducted by the building 

department staff, by volunteer engineers and architects, by private consultants 

retained by individual building owners, or by a combination of these. Due to the 

limited resources available, it is usually necessary to limit these postearthquake 

inspections to those structures most likely to have been severely damaged and to 

make a rapid assessment of the severity of damage. 

Following initial postearthquake assessment, buildings are typically tagged with 

a placard to inform the owner and public of the assessed condition. “Green 

tags” are typically used to indicate that the building has been subjected to a rapid 

inspection and does not appear to have sustained damage that impairs its safety 

for occupancy. “Yellow tags” are typically used to indicate a condition of 

limited, or perhaps unknown, impairment of building safety. “Red tags” are 

commonly used to indicate that a building has been assessed as unsafe for further 

occupancy. Once a building has been posted with either a yellow or red tag, the 

building owner must take action to clear this posting. Typically the owner must 

retain a consultant to perform more detailed inspections and evaluations, and 

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either report back to the building official that the building was not seriously 

damaged, or to prepare recommendations for repair of the structure and to have 

the posting removed. Note that only the building official is authorized to allow the 

posting to be altered or removed. 

These Recommended Criteria provide guidelines for performing the rapid postearthquake assessments, typically conducted by the building official; for 

performing the more detailed assessments, typically performed by a private 

consultant under contract to the building owner; and for developing repair 

programs. These repair programs are intended to restore the structure to the 

approximate condition and level of safety that existed prior to the onset of 

damage in this particular earthquake event. These Recommended Criteria do not 

specifically provide recommendations for upgrade of a building, to improve its 

performance in the event of future earthquake ground shaking. 

In many cases, when a building experiences severe damage in a relatively 

moderate event, this damage is an indication that the building is vulnerable and 

could experience more extensive and severe damage in future events. In 

recognition of this, many locally adopted building codes contain provisions that 

require upgrade of structures, as well as repair, when they have been damaged 

beyond a certain level. This “trigger” level for upgrade varies widely from 

community to community. Regardless of whether or not the local building code 

requires upgrade as well as repair, an upgrade should be considered by the 

owner at the time structural repairs are conducted. For technical criteria for 

evaluating the advisability of upgrades, and methods of designing such upgrades, 

refer to FEMA-351, Recommended Seismic Evaluation and Upgrade Criteria for 

Existing Welded Steel Moment-Frame Buildings. 

When a decision is made to repair a structure, without upgrade, the engineer is 

cautioned to alert the owner that similar or perhaps more severe damage could be 

anticipated in future events. Further, the engineer should take care that in the 

process of conducting repairs, conditions of structural irregularity, discontinuity, 

or strength or stiffness deficiency are not introduced into the structure, and that 

existing such conditions are not made more severe. 

1.3 Background 

For many years, the basic intent of the building code seismic provisions has been to provide 

buildings with an ability to withstand intense ground shaking without collapse, but potentially 

with some significant structural damage. In order to accomplish this, one of the basic principles 

inherent in modern code provisions is to encourage the use of building configurations, structural 

systems, materials and details that are capable of ductile behavior. A structure is said to behave 

in a ductile manner if it is capable of withstanding large inelastic deformations without 

significant degradation in strength, and without the development of instability and collapse. The 

design forces specified by building codes for particular structural systems are related to the 

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Steel Moment-Frame Buildings Chapter 1: Introduction 

amount of ductility the system is deemed to possess. Generally, structural systems with more 

ductility are designed for lower forces than less ductile systems, as ductile systems are deemed 

capable of resisting demands that are significantly greater than their elastic strength limit. 

Starting in the 1960s, engineers began to regard welded steel moment-frame buildings as being 

among the most ductile systems contained in the building code. Many engineers believed that 

steel moment-frame buildings were essentially invulnerable to earthquake-induced structural 

damage and thought that should such damage occur, it would be limited to ductile yielding of 

members and connections. Earthquake-induced collapse was not believed possible. Partly as a 

result of this belief, many large industrial, commercial and institutional structures employing 

steel moment-frame systems were constructed, particularly in the western United States. 

The Northridge earthquake of January 17, 1994 challenged this paradigm. Following that 

earthquake, a number of steel moment-frame buildings were found to have experienced brittle 

fractures of beam-to-column connections. The damaged buildings had heights ranging from one 

story to 26 stories, and a range of ages spanning from buildings as old as 30 years to structures 

being erected at the time of the earthquake. The damaged buildings were spread over a large 

geographical area, including sites that experienced only moderate levels of ground shaking. 

Although relatively few buildings were located on sites that experienced the strongest ground 

shaking, damage to buildings on these sites was extensive. Discovery of these unanticipated 

brittle fractures of framing connections, often with little associated architectural damage to the 

buildings, was alarming to engineers and the building industry. The discovery also caused some 

concern that similar, but undiscovered, damage may have occurred in other buildings affected by 

past earthquakes. Later investigations confirmed such damage in a limited number of buildings 

affected by the 1992 Landers, 1992 Big Bear and 1989 Loma Prieta earthquakes. 

In general, steel moment-frame buildings damaged by the 1994 Northridge earthquake met 

the basic intent of the building codes. That is, they experienced limited structural damage, but 

did not collapse. However, the structures did not behave as anticipated and significant economic 

losses occurred as a result of the connection damage, in some cases, in buildings that had 

experienced ground shaking less severe than the design level. These losses included direct costs 

associated with the investigation and repair of this damage as well as indirect losses relating to 

the temporary, and in a few cases, long-term, loss of use of space within damaged buildings. 

Steel moment-frame buildings are designed to resist earthquake ground shaking based on the 

assumption that they are capable of extensive yielding and plastic deformation, without loss of 

strength. The intended plastic deformation consists of plastic rotations developing within the 

beams, at their connections to the columns, and is theoretically capable of resulting in benign 

dissipation of the earthquake energy delivered to the building. Damage is expected to consist of 

moderate yielding and localized buckling of the steel elements, not brittle fractures. Based on this 

presumed behavior, building codes permit steel moment-frame buildings to be designed with a 

fraction of the strength that would be required to respond to design level earthquake ground shaking 

in an elastic manner. 

Steel moment-frame buildings are anticipated to develop their ductility through the 

development of yielding in beam-column assemblies at the beam-column connections. This 

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yielding may take the form of plastic hinging in the beams (or less desirably, in the columns), 

plastic shear deformation in the column panel zones, or through a combination of these 

mechanisms. It was believed that the typical connection employed in steel moment-frame 

construction, shown in Figure 1-1, was capable of developing large plastic rotations, on the order 

of 0.02 radians or larger, without significant strength degradation. 

Observation of damage sustained by buildings in the 1994 Northridge earthquake indicated 

that contrary to the intended behavior, in many cases brittle fractures initiated within the 

connections at very low levels of plastic demand, and in some cases, while the structures 

remained essentially elastic. Typically, but not always, fractures initiated at the complete joint 

penetration (CJP) weld between the beam bottom flange and column flange (Figure 1-2). Once 

initiated, these fractures progressed along a number of different paths, depending on the 

individual joint conditions. 

Figure 1-1 Typical Welded Moment-Resisting Connection Prior to 1994 

In some cases, the fractures progressed completely through the thickness of the weld, and 

when fire protective finishes were removed, the fractures were evident as a crack through 

exposed faces of the weld, or the metal just behind the weld (Figure 1-3a). Other fracture 

patterns also developed. In some cases, the fracture developed into a crack of the column flange 

material behind the CJP weld (Figure 1-3b). In these cases, a portion of the column flange 

remained bonded to the beam flange, but pulled free from the remainder of the column. This 

fracture pattern has sometimes been termed a “divot” or “nugget” failure. 

A number of fractures progressed completely through the column flange, along a nearhorizontal plane that aligns approximately with the beam lower flange (Figure 1-4a). In some 

cases, these fractures extended into the column web and progressed across the panel zone (Figure 

1-4b). Investigators have reported some instances where columns fractured entirely across the 

section. 

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Backing bar Backing bar

Beam flange Beam flange

Fracture Fracture

Backing bar

Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings Chapter 1: Introduction 

Beam flange 

Backing bar 

Fracture 

Figure 1-2 Common Zone of Fracture Initiation in Beam-Column Connection 

a. Fracture at Fused Zone b. Column Flange "Divot" Fracture 

Figure 1-3 Fractures of Beam to Column Joints 

a. Fractures through Column Flange b. Fracture Progresses into Column Web 

Figure 1-4 Column Fractures 

Once such fractures have occurred, the beam-column connection has experienced a 

significant loss of flexural rigidity and strength to resist those loads that tend to open the crack. 

Residual flexural strength and rigidity must be developed through a couple consisting of forces 

transmitted through the remaining top flange connection and the web bolts. However, in 

Column flange Column flange

Fused zone Fused zone

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providing this residual strength and stiffness, the bolted web connections can themselves be 

subject to failures. These include fracturing of the welds of the shear plate to the column, 

fracturing of supplemental welds to the beam web or fracturing through the weak section of shear 

plate aligning with the bolt holes (Figure 1-5). 

Despite the obvious local strength impairment resulting from these fractures, many damaged 

buildings did not display overt signs of structural damage, such as permanent drifts or damage to 

architectural elements, making reliable postearthquake damage evaluations difficult. In order to 

determine reliably if a building has sustained connection damage it is necessary to remove 

architectural finishes and fireproofing, and perform detailed inspections of the connections. 

Even if no damage is found, this is a costly process. Repair of damaged connections is even 

more costly. At least one steel moment-frame building sustained so much damage that it was 

deemed more practical to demolish the building than to repair it. 

Figure 1-5 Vertical Fracture through Beam Shear Plate Connection 

Initially, the steel construction industry took the lead in investigating the causes of this 

unanticipated damage and in developing design recommendations. The American Institute of 

Steel Construction (AISC) convened a special task committee in March, 1994 to collect and 

disseminate available information on the extent of the problem (AISC, 1994a). In addition, 

together with a private party engaged in the construction of a major steel building at the time of 

the earthquake, AISC participated in sponsoring a limited series of tests of alternative connection 

details at the University of Texas at Austin (AISC, 1994b). The American Welding Society 

(AWS) also convened a special task group to investigate the extent to which the damage was 

related to welding practice, and to determine if changes to the welding code were appropriate 

(AWS, 1995). 

In September, 1994, the SAC Joint Venture, AISC, the American Iron and Steel Institute and 

National Institute of Standards and Technology jointly convened an international workshop 

(SAC, 1994) in Los Angeles to coordinate the efforts of the various participants and to lay the 

foundation for systematic investigation and resolution of the problem. Following this workshop, 

FEMA entered into a cooperative agreement with the SAC Joint Venture to perform problemfocused studies of the seismic performance of steel moment-frame buildings and to develop 

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recommendations for professional practice (Phase I of SAC Steel Project). Specifically, these 

recommendations were intended to address the following: the inspection of earthquake-affected 

buildings to determine if they had sustained significant damage; the repair of damaged buildings; 

the upgrade of existing buildings to improve their probable future performance; and the design of 

new structures to provide reliable seismic performance. 

During the first half of 1995, an intensive program of research was conducted to explore 

more definitively the pertinent issues. This research included literature surveys, data collection 

on affected structures, statistical evaluation of the collected data, analytical studies of damaged 

and undamaged buildings, and laboratory testing of a series of full-scale beam-column 

assemblies representing typical pre-Northridge design and construction practice as well as 

various repair, upgrade and alternative design details. The findings of these tasks formed the 

basis for the development of FEMA-267 – Interim Guidelines: Evaluation, Repair, Modification, 

and Design of Welded Steel Moment Frame Structures, which was published in August, 1995. 

FEMA-267 provided the first definitive, albeit interim, recommendations for practice, following 

the discovery of connection damage in the 1994 Northridge earthquake. 

In September 1995 the SAC Joint Venture entered into a contractual agreement with FEMA 

to conduct Phase II of the SAC Steel Project. Under Phase II, SAC continued its extensive 

problem-focused study of the performance of moment resisting steel frames and connections of 

various configurations, with the ultimate goal of develop seismic design criteria for steel 

construction. This work has included: extensive analyses of buildings; detailed finite element 

and fracture mechanics investigations of various connections to identify the effects of connection 

configuration, material strength, and toughness and weld joint quality on connection behavior; as 

well as more than 120 full-scale tests of connection assemblies. As a result of these studies, and 

independent research conducted by others, it is now known that the typical moment-resisting 

connection detail employed in steel moment-frame construction prior to the 1994 Northridge 

earthquake, and depicted in Figure 1-1, had a number of features that rendered it inherently 

susceptible to brittle fracture. These included the following: 

• The most severe stresses in the connection assembly occur where the beam joins to the 

column. Unfortunately, this is also the weakest location in the assembly. At this location, 

bending moments and shear forces in the beam must be transferred to the column through the 

combined action of the welded joints between the beam flanges and column flanges and the 

shear tab. The combined section properties of these elements, for example the cross sectional 

area and section modulus, are typically less than those of the connected beam. As a result, 

stresses are locally intensified at this location. 

• The joint between the bottom beam flange and the column flange is typically made as a 

downhand field weld, often by a welder sitting on top of the beam top flange, in a so-called 

“wildcat” position. To make the weld from this position each pass must be interrupted at the 

beam web, with either a start or stop of the weld at this location. This welding technique 

often results in poor quality welding at this critical location, with slag inclusions, lack of 

fusion and other defects. These defects can serve as crack initiators, when the connection is 

subjected to severe stress and strain demands. 

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Chapter 1: Introduction Steel Moment-Frame Buildings 

• The basic configuration of the connection makes it difficult to detect hidden defects at the 

root of the welded beam-flange-to-column-flange joints. The backing bar, which was 

typically left in place following weld completion, restricts visual observation of the weld 

root. Therefore, the primary method of detecting defects in these joints is through the use of 

ultrasonic testing (UT). However, the geometry of the connection also makes it very difficult 

for UT to detect flaws reliably at the bottom beam flange weld root, particularly at the center 

of the joint, at the beam web. As a result, many of these welded joints have undetected 

significant defects that can serve as crack initiators. 

• Although typical design models for this connection assume that nearly all beam flexural 

stresses are transmitted by the flanges and all beam shear forces by the web, in reality, due to 

boundary conditions imposed by column deformations, the beam flanges at the connection 

carry a significant amount of the beam shear. This results in significant flexural stresses on 

the beam flange at the face of the column, and also induces large secondary stresses in the 

welded joint. Some of the earliest investigations of these stress concentration effects in the 

welded joint were conducted by Richard, et al. (1995). The stress concentrations resulting 

from this effect resulted in severe strength demands at the root of the complete joint 

penetration welds between the beam flanges and column flanges, a region that often includes 

significant discontinuities and slag inclusions, which are ready crack initiators. 

• In order that the welding of the beam flanges to the column flanges be continuous across the 

thickness of the beam web, this detail incorporates weld access holes in the beam web, at the 

beam flanges. Depending on their geometry, severe strain concentrations can occur in the 

beam flange at the toe of these weld access holes. These strain concentrations can result in 

low-cycle fatigue and the initiation of ductile tearing of the beam flanges after only a few 

cycles of moderate plastic deformation. Under large plastic flexural demands, these ductile 

tears can quickly become unstable and propagate across the beam flange. 

• Steel material at the center of the beam-flange-to-column-flange joint is restrained from 

movement, particularly in connections of heavy sections with thick column flanges. This 

condition of restraint inhibits the development of yielding at this location, resulting in locally 

high stresses on the welded joint, which exacerbates the tendency to initiate fractures at 

defects in the welded joints. 

• Design practice in the period 1985-1994 encouraged design of these connections with 

relatively weak panel zones. In connections with excessively weak panel zones, inelastic 

behavior of the assembly is dominated by shear deformation of the panel zone. This panel 

zone shear deformation results in a local kinking of the column flanges adjacent to the beamflange-to-column-flange joint, and further increases the stress and strain demands in this 

sensitive region. 

In addition to the above, additional conditions contributed significantly to the vulnerability of 

connections constructed prior to 1994. 

• In the mid-1960s, the construction industry moved to the use of the semi-automatic, selfshielded, flux-cored arc welding process (FCAW-S) for making the joints of these 

connections. The welding consumables that building erectors most commonly used 

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inherently produced welds with very low toughness. The toughness of this material could be 

further compromised by excessive deposition rates, which unfortunately were commonly 

employed by welders. As a result, brittle fractures could initiate in welds with large defects, 

at stresses approximating the yield strength of the beam steel, precluding the development of 

ductile behavior. 

• Early steel moment frames tended to be highly redundant and nearly every beam-column joint 

was constructed to behave as part of the lateral-force-resisting system. As a result, member 

sizes in these early frames were small and much of the early acceptance testing of this typical 

detail were conducted with specimens constructed of small framing members. As the cost of 

construction labor increased, the industry found that it was more economical to construct 

steel moment-frame buildings by moment-connecting a relatively small percentage of the 

beams and columns and by using larger members for these few moment-connected elements. 

The amount of strain demand placed on the connection elements of a steel moment frame is 

related to the span-to-depth ratio of the member. Therefore, as member sizes increased, 

strain demands on the welded connections also increased, making the connections more 

susceptible to brittle behavior. 

• In the 1960s and 1970s, when much of the initial research on steel moment-frame 

construction was performed, beams were commonly fabricated using A36 material. In the 

1980s, many steel mills adopted more modern production processes, including the use of 

scrap-based production. Steels produced by these more modern processes tended to include 

micro-alloying elements that increased the strength of the materials so that despite the 

common specification of A36 material for beams, many beams actually had yield strengths 

that approximated or exceeded that required for grade 50 material. As a result of this 

increase in base metal yield strength, the weld metal in the beam-flange-to-column-flange 

joints became under-matched, potentially contributing to its vulnerability. 

At this time, it is clear that in order to obtain reliable ductile behavior of steel moment-frame 

construction a number of changes to past practices in design, materials, fabrication, erection and 

quality assurance are necessary. The recommended criteria contained in this document, and the 

companion publications, are based on an extensive program of research into materials, welding 

technology, inspection methods, frame system behavior, and laboratory and analytical 

investigations of different connection details. The recommended criteria presented herein are 

believed to be capable of addressing the vulnerabilities identified above and providing for frames 

capable of more reliable performance in response to earthquake ground shaking. 

Although many of the above conditions developed incrementally, over a period of twenty or 

more years, most steel moment-frame buildings constructed during the period 1960-1994 

employed connections of a type that is subject to these vulnerabilities. Therefore, all steel 

moment-frame buildings constructed during this period should be considered vulnerable to 

brittle, earthquake-induced, connection fractures, unless specific evidence is available that 

indicates these vulnerabilities are not present. 

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Chapter 1: Introduction Steel Moment-Frame Buildings 

Commentary: The typical moment connection detail employed in most welded 

steel moment-frames constructed during the period 1960-1994 is that shown in 

Figure 1-1. Although the properties of structural steels and weld metals 

employed in fabricating and constructing these connections varied somewhat over 

the years, the basic configuration was almost universally applied in this 

construction type during this time period. It is now known that almost all such 

connections can be subject to fracture at levels of inelastic demand that are 

significantly below those currently believed to be appropriate. Therefore, 

following strong earthquake ground shaking, unless suitable evidence is available 

to indicate that a building does not have vulnerable connections, or if evaluations 

conducted in accordance with these recommendations indicate that significant 

damage is unlikely to have occurred, all steel moment-frame buildings 

constructed during the period 1960-1994 should be considered to be potentially 

damaged. Suitable evidence that a building does not have vulnerable connections 

could include original construction documents that portray connection details 

that are substantially different from those indicated in Figure 1-1. 

1.4 Application 

These Recommended Criteria supersede the postearthquake evaluation and repair guidelines 

for existing steel moment-frame buildings contained in FEMA-267, Interim Guidelines: 

Evaluation, Repair, Modification and Design of Welded Steel Moment Frame Structures, and the 

Interim Guidelines Advisories, Nos. 1 and 2 (FEMA-267A and FEMA-267B). This document has 

been prepared in coordination with FEMA-302 1997 NEHRP Recommended Provisions for 

Seismic Regulations for New Buildings and Other Structures, the 1997 AISC Seismic 

Specification (AISC, 1997) and the 1998 AWS D1.1 Structural Welding Code – Steel (AWS, 

1998). Users are cautioned to consider carefully any differences between the aforementioned 

documents and those actually enforced by the building department having jurisdiction for a 

specific project and to adjust the recommendations contained here accordingly. 

1.5 Postearthquake Evaluation and Repair Process 

Postearthquake evaluation of a welded steel moment-frame is a multi-step process (Figure 

1-6). The intent is to identify buildings that have sustained sufficient structural damage to 

compromise future performance, determine the extent and severity of this damage, assess the 

general implications of the damage with regard to building safety and determine appropriate 

actions regarding building occupancy and repair. Once a determination is made that a building 

has sustained significant damage the structural engineer should conduct a more detailed 

evaluation of the structure’s residual structural integrity and safety and develop a detailed plan 

for repair, upgrade, demolition, or other action, as appropriate. 

Currently, most building codes only require repair of damaged structures, not upgrade. As 

such, the focus of this document is the identification and repair of damage. However, the extent, 

severity or characteristics of damage may be sufficiently severe that the owner may wish to 

consider upgrading or modifying the structure to improve probable performance in future events. 

Such action may be particularly appropriate when a building has sustained severe damage as a 

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Damaging event occurs 

Preliminary 

Evaluation 

(Chapter 3) 

Ground Motion 

Screening 

(Chapter 3) 

Ground Motion 

Exceed 

Triggers? 

yes 

no No 

Evaluation or Posting 

Required 

Post 

Building 

Green, Yellow or Red 

Detailed

Inspection

(Chapter 4)

Report 

Detailed 

Evaluation 

Level 1 

(Chapter 4) 

Level 2 

(Chapter 5) 

Revise Posting, 

Repair (Chapter 6) 

or Upgrade (FEMA−351) 

Figure 1-6 Flow Chart for Postearthquake Actions 

result of moderate ground shaking, as this may indicate an inability to reliably resist failure in 

stronger events. Prediction of structural performance during future earthquakes and selection of 

appropriate upgrades to achieve desired performance is the subject of a companion document, 

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Chapter 1: Introduction Steel Moment-Frame Buildings 

FEMA-351 – Recommended Seismic Evaluation and Upgrade Criteria for Welded Steel MomentFrame Buildings. 

The first step in the evaluation process is a screening to identify those buildings unlikely to 

have experienced ground motion of sufficient intensity to cause significant damage. Since strong 

ground motion instruments are installed in relatively few buildings, it is typically necessary to 

estimate the regional distribution of ground motion intensity using available instrumental 

recordings and observed patterns of damage. Those buildings suspected of having experienced 

ground motion of sufficient intensity to cause damage should be subjected to a rapid on-site 

evaluation, to determine if there are obvious indications of potentially life threatening conditions. 

Following this rapid evaluation, the building should be posted, to indicate whether such 

conditions were found. Criteria for performing the initial screening and rapid on-site evaluations 

are presented in Chapter 3. 

Often, damage to steel moment-frame buildings cannot be detected by rapid evaluations like 

those presented in Chapter 3. Therefore, buildings suspected of having experienced potentially 

damaging ground motion should also be subjected to more detailed inspections and evaluation. 

Chapter 4 outlines a simplified method for such evaluations, similar to that contained in FEMA267. Chapter 5 presents an alternative, more rigorous procedure consistent with that used for 

structural performance assessments in other documents prepared by the FEMA/SAC project. 

Both of these procedures contain recommendations for inspection of some or all steel momentframe connections in the building; classification of the damage found (in accordance with a 

system presented in Chapter 2); assessment of the safety of the building, and development of 

recommendations for repair or other remedial action. Methods of conducting repair and criteria 

for specifying these methods are presented in Chapter 6. These recommendations do not cover 

routine correction of non-conforming conditions resulting from deficiencies in the original 

construction. Industry standard practices are acceptable for such repairs. Recommended criteria 

for the assessment of seismic performance of the repaired building and recommendations for 

improved performance may be found in the companion publication, FEMA-351. 

1.6 Overview of These Recommended Criteria 

The following is an overview of the general contents of the chapters contained in these 

Recommended Criteria, and their intended use: 

• Chapter 2: Inspection and Classification of Damage. This chapter provides an overview 

of the different types of structural damage that may be anticipated to occur in welded steel 

moment-frame buildings, together with a discussion of their significance. This chapter also 

introduces a damage classification system that is referenced throughout the remaining 

chapters. 

• Chapter 3: Preliminary Postearthquake Assessment. This chapter provides screening 

criteria that can be used to determine if there is sufficient likelihood that a welded steel 

moment-resisting frame structure has experienced significant damage to warrant further 

investigation. This Chapter also provides a preliminary evaluation procedure that may be 

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Steel Moment-Frame Buildings Chapter 1: Introduction 

rapidly performed to determine if the building presents imminent safety hazards. Building 

officials may use the screening criteria to determine which buildings should be subjected to 

inspections by the Building Department using the Preliminary Evaluation Procedures. While 

these preliminary evaluation procedures should permit the identification of structures with 

damage so severe that imminent hazards have been created, they will typically not be 

sufficient to determine if more moderate levels of damage have occurred. Chapters 4 and 5 

provide procedures for more detailed evaluations, necessary to make such determination. 

• Chapter 4: Level 1 Detailed Postearthquake Evaluations. Except for those structures that 

have experienced partial or total collapse, or that exhibit significant permanent interstory 

drift, the results of a preliminary evaluation conducted in accordance with Chapter 3 are 

likely to be inconclusive with regard to the postearthquake condition of the structure. This 

chapter provides procedures for conducting more detailed evaluations of the building to 

confirm its postearthquake condition and develop recommendations for occupancy and repair 

of the structure as appropriate. It includes performing inspections of the fracture-susceptible 

connections in the structure, to determine their condition, and calculation of a damage index. 

Recommendations for occupancy restriction and repair are provided, based on the calculated 

value of the damage index. This level of evaluation is too lengthy to be conducted as part of 

the rapid postearthquake assessments typically conducted by building departments and is 

anticipated to be implemented by engineers engaged by the building owner. 

• Chapter 5: Level 2 Detailed Postearthquake Evaluations. If a building has experienced 

many connection fractures, and other types of structural damage, as revealed by a level 1, 

detailed evaluation, then it may be advisable to restrict occupancy of the building until it can 

be repaired. Decisions to restrict occupancy can result in a large economic burden, both for 

the building owner and the tenants and some engineers may be reluctant to advise such action 

unless analytical evaluation indicates the presence of significant safety hazards. This chapter 

provides an analytical methodology for estimating the probability of earthquake-induced 

collapse of a damaged building that can be used to supplement occupancy decisions 

suggested by the evaluation procedures of Chapter 4. 

• Chapter 6: Postearthquake Repair. This chapter provides recommendations for repair of 

the most common types of damage encountered in welded steel moment-frame construction. 

It does not include guidelines for structural upgrade. Often, the most logical time to conduct 

a structural upgrade is during the time that earthquake damage is being repaired. In addition, 

some jurisdictions require upgrade of buildings that have sustained extensive damage as a 

matter of policy. Criteria for performing structural upgrade may be found in a companion 

publication, FEMA-351 – Recommended Seismic Evaluation and Upgrade Criteria for 

Existing Welded Steel Moment-Frame Buildings. 

• Appendix A: Detailed Procedures for Performance Evaluation. This appendix describes 

in detail the basis of the reliability-based evaluation methods presented in Chapter 5. It may 

be used to obtain more certain estimates of structural capacity and must be used for that 

purpose, instead of the procedures of Chapter 5, for irregular structures. 

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Chapter 1: Introduction Steel Moment-Frame Buildings 

• Appendix B: Sample Placards. This appendix contains sample placards that may be used 

to post buildings following preliminary postearthquake evaluations conducted in accordance 

with Chapter 3 (from ATC, 1995). 

• Appendix C: Sample Inspection Forms. This appendix contains a series of forms that may 

be used to record damage detected in beam-column connections as part of a detailed 

postearthquake inspection program conducted in accordance with Chapter 4. 

• References, Bibliography, and Acronyms. 

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2. INSPECTION AND CLASSIFICATION OF DAMAGE 

2.1 Introduction 

This chapter defines a uniform system for classification and reporting of damage to steel 

moment-frame structures that have been subjected to strong earthquake ground shaking. 

Structural damage observed in steel moment-frame buildings following strong ground 

shaking can include yielding, buckling and fracturing of the steel framing elements (beams and 

columns) and their connections, as well as permanent lateral drift. Damaged elements can 

include girders, columns, column panel zones (including girder flange continuity plates and 

column web doubler plates), the welds of the beam to column flanges, the shear tabs which 

connect the girder webs to column flanges, column splices and base plates. Figure 2-1 illustrates 

the location of these elements. 

Frame Elevation 

Column splice 

Girder 

Base Plate 

Column 

Doubler Plate 

Continuity Plate 

Weld 

Panel Zone 

Shear Tab 

Figure 2-1 Elements of Welded Steel Moment Frame 

2.2 Damage Types 

Damage to framing elements of steel moment-frame buildings may be categorized as 

belonging to the weld (W), girder (G), column (C), panel zone (P) or shear tab (S) categories. 

This section defines a uniform system for classification and reporting of damage to elements of 

steel moment-frame structures that is utilized throughout these Recommended Criteria. The 

damage types indicated below are not mutually exclusive. A given girder-column connection, for 

example, may exhibit several different types of damage. In addition to the individual element 

damage types, a damaged steel moment-frame may also exhibit global effects, such as permanent 

interstory drifts. 

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Chapter 2: Inspection and Classification of Damage Steel Moment-Frame Buildings 

Following a detailed postearthquake inspection, classification of the damage found, as to its 

type and degree of severity, is the first step in performing an assessment of the condition and 

safety of a damaged steel moment-frame structure. In a level 1 evaluation, conducted in 

accordance with Chapter 4 of these Recommended Criteria, the classifications of this section are 

used for the assignment of damage indices. These damage indices are statistically combined and 

extrapolated to provide an indication of the severity of damage to a structure's lateral force 

resisting system and are used as a basis for selecting building repair strategies. For a level 2 

evaluation, conducted in accordance with Chapter 5 of these Recommended Criteria, these 

damage classifications are keyed to specific modeling recommendations for analysis of damaged 

buildings to determine their response to likely ground shaking in the immediate postearthquake 

period. Chapter 6 addresses specific techniques and design criteria recommended for the repair 

and modification of the different types of damage, keyed to these same damage classifications. 

Commentary: The damage types contained in this chapter are based on a system 

first defined in a statistical study of damage reported in NISTR-5625 (Youssef et 

al., 1995). The original classes contained in that study have been expanded 

somewhat to include some conditions not previously identified. 

2.2.1 Girder Damage 

Girder damage may consist of yielding, buckling or fracturing of the flanges of girders at or 

near the girder-column connection. Seven separate types are defined in Table 2-1. Figure 2-2 

illustrates these various types of damage. See Section 2.2.3 and 2.2.4 for damage to adjacent 

welds and shear tabs, respectively. 

Commentary: Minor yielding of girder flanges (type G2) is the least significant 

type of girder damage. It is often difficult to detect and may be exhibited only by 

local flaking of mill scale and the formation of characteristic visible lines in the 

material, running across the flange. Removal of finishes, by scraping, may often 

obscure the detection of this type of damage. Girder flange yielding, without 

local buckling or fracture, results in negligible degradation of structural strength 

and typically need not be repaired. 

Table 2-1 Types of Girder Damage 

Type Description 

G1 Buckled flange (top or bottom) 

G2 Yielded flange (top or bottom) 

G3 Flange fracture in Heat Affected Zone (top or bottom) 

G4 Flange fracture outside Heat Affected Zone (top or 

bottom) 

G5 Not used 

G6 Yielding or buckling of web 

G7 Fracture of web 

G8 Lateral torsion buckling of section 

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G1 

G3 G2

G4 

G7 

G6 

G8

Figure 2-2 Types of Girder Damage 

Girder flange buckling (type G1) can result in a significant loss of girder 

plastic strength, particularly when accompanied by girder web buckling (type 

G6). For compact sections, this strength loss occurs gradually, and increases 

with the number of inelastic cycles and the extent of the inelastic excursion. 

Following the initial onset of buckling, additional buckling will often occur at 

lower load levels and result in further reductions in strength, compared to 

previous cycles. The localized secondary stresses which occur in the girder 

flanges due to the buckling can result in initiation of flange fracture damage (G4) 

if the frame is subjected to a large number of cycles. Such fractures typically 

progress slowly over repeated cycles, and grow in a ductile manner. Once this 

type of damage initiates, the girder flange will begin to lose tensile capacity 

under continued or reversed loading, although it may retain some capacity in 

compression. Visually evident girder flange buckling should be repaired. 

In structures with weld material with low notch-toughnesss, girder flange 

cracking within the Heat Affected Zone (HAZ) (type G3) can occur as an 

extension of brittle fractures that initiate in the weld root. This is particularly 

likely to occur at connections in which improper welding procedures were 

followed, resulting in a brittle HAZ. However, these fractures can also occur in 

connections with welded joints made with notch-tough weld metal and following 

appropriate procedures, as a result of low-cycle fatigue, exacerbated by the very 

high strain demands that occur at the toe of the weld access hole, in unreinforced 

beam-column connections. Like the visually similar type G4 damage, which can 

also result from low cycle fatigue conditions at the toe of the weld access hole, it 

results in a complete loss of flange tensile capacity, and consequently, significant 

reduction in the contribution to frame lateral strength and stiffness from the 

connection. 

In the 1994 Northridge earthquake girder damage was most commonly 

detected at the bottom flanges, although some instances of top flange failure were 

also reported. There are several reasons for this. First, the composite action 

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induced by the presence of a floor slab at the girder top flange tends to shift the 

neutral axis of the beam towards the top flange. This results in larger tensile 

deformation demands on the bottom flange than on the top. In addition, the 

presence of the slab tends to reduce the chance of local buckling of the top flange. 

The bottom flange being less restrained can experience buckling relatively easily. 

Finally, much of the damage found in girders initiates as a result of defects at the 

root of the beam flange to column flange weld. Due to its position, the weld of the 

bottom beam flange to column flange is more difficult to make than that at the top 

flange, and therefore, is more likely to have defects that can initiate such damage. 

2.2.2 Column Flange Damage 

Seven types of column flange damage are defined in Table 2-2 and illustrated in Figure 2-3. 

Column flange damage typically results in degradation of a structure's gravity-load-carrying 

strength as well as lateral-load resistance. For related damage to column panel zones, refer to 

Section 2.2.5. 

Table 2-2 Types of Column Damage 

Type Description 

C1 Minor column flange surface crack 

C2 Flange tear-out or divot 

C3 Full or partial flange crack outside Heat Affected Zone 

C4 Full or partial flange crack in Heat Affected Zone 

C5 Lamellar flange tearing 

C6 Buckled flange 

C7 Column splice failure 

C1 

C2 

C3 

C4 

C5 

C6 

C7 

Figure 2-3 Types of Column Damage 

Commentary: Column flange damage includes types C1 through C7. Type C1 

damage consists of a small crack at the surface of the column flange and 

extending into its thickness, typically at the location of the adjoining girder 

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flange. C1 damage does not go through the thickness of the column flange and 

can often be detected only by nondestructive testing (NDT). Type C2 damage is 

an extension of type C1, in which a curved failure surface extends from an 

initiation point, usually at the root of the girder-to-column-flange weld, and 

extends longitudinally into the column flange. In some cases this failure surface 

may emerge on the same face of the column flange as the one where it initiated. 

When this occurs, a characteristic "nugget" or "divot" can be withdrawn from the 

flange. Types C3 and C4 fractures extend through the thickness of the column 

flange and may extend into the panel zone. Type C5 damage is characterized by a 

step-shaped failure surface within the thickness of the column flange and aligned 

parallel to it. This damage is often detectable only with the use of nondestructive 

testing. 

Type C1 damage does not result in an immediate large strength loss in the 

column; however, such small fractures can easily progress into more serious 

types of damage if subjected to additional large tensile loading by aftershocks or 

future earthquakes. Type C2 damage may result in both a loss of effective 

attachment of the girder flange to the column for tensile demands and could cause 

a significant reduction in available column flange area for resistance of axial and 

flexural demands. Type C3 and C4 damage result in a loss of column flange 

tensile capacity and under additional loading can progress into other types of 

damage. 

Type C5 damage may occur as a result of non-metallic inclusions within the 

column flange. The potential for this type of fracture under conditions of high 

restraint and large through-thickness tensile demands, such as the residual 

stresses induced by welding, has been known for a number of years, and is termed 

lamellar tearing. There is no evidence that lamellar tearing actually occurred in 

buildings as a result of earthquake ground shaking and it is currently thought that 

when type C5 damage did occur, it was an extension of fracturing that initiated in 

the weld root. This damage has sometimes been identified as a potential 

contributing mechanism for type C2 column flange through-thickness failures. 

Note that in many cases, type C2 damage may be practically indistinguishable 

from type W3 fractures (see Section 2.2.3). The primary difference is that in type 

W3, the fracture surface generally remains with in the heat affected zone of the 

column flange material while in C2 damage, the fracture surface progresses 

deeper into the column flange material. 

Type C6 damage consists of local buckling of the column flange, adjacent to 

the beam-column connection. While such damage was not actually observed in 

buildings following the 1994 Northridge earthquake, it can be anticipated at 

locations where plastic hinges form in the columns. Buckling of beam flanges has 

been observed in the laboratory at interstory drift demands in excess of 0.02 

radians. Column sections are usually more compact than beams and therefore, 

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are less prone to local buckling. Type C6 damage may occur, however, in 

buildings with strong-beam-weak-column systems and at the bases of columns in 

any building when very large interstory drifts have occurred. 

Type C7 damage, fracturing of welded column splices, also was not observed 

following the Northridge earthquake. However, the partial joint penetration 

groove welds commonly used in these splices are very susceptible to fracture 

when subjected to large tensile loads. Large tensile loads can occur on a column 

splice as a result of global overturning effects, or as a result of large flexural 

demands in the column. 

As a result of the potential safety consequences of complete column failure, all 

column damage should be considered as significant and repaired expeditiously. 

2.2.3 Weld Damage 

Three types of weld damage are defined in Table 2-3 and illustrated in Figure 2-4. All apply 

to the complete joint penetration welds between the girder flanges and the column flanges. 

Table 2-3  Types of Weld Damage, Defects and Discontinuities 

Type  Description 

W1, W1a, W1b Not Used (see commentary) 

W2 Crack through weld metal thickness 

W3 Fracture at column interface 

W4 Fracture at girder flange interface 

W5 Not Used (see commentary) 

W3 W4 

W2 

Figure 2-4  Types of Weld Damage 

Commentary: In addition to the W2, W3, and W4 types of damage indicated in 

Table 2-3 and Figure 2-4, the damage classification system presented in FEMA267 included conditions at the root of the complete joint penetration weld that did 

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not propagate through the weld nor into the surrounding base metal, and could be 

detected only by removal of the weld backing or through the use of nondestructive 

testing (NDT). These conditions were termed types W1a, W1b, and W5. 

As defined in FEMA-267, type W5 consisted of small discontinuities at the 

root of the weld, which, if discovered as part of a construction quality control 

program for new construction, would not be rejectable under the AWS D1.1 

provisions. FEMA-267 recognized that W5 conditions were likely to be the result 

of acceptable flaws introduced during the initial building construction, but 

included this classification so that such conditions could be reported in the event 

they were detected in the course of the ultrasonic testing (UT) that FEMA-267 

required. There was no requirement to repair such conditions. Since these 

Recommended Criteria do not require UT as a routine part of the inspection 

protocol, W5 conditions are unlikely to be detected and have been omitted as a 

damage classification. 

Type W1a and W1b conditions, as contained in FEMA-267, consisted of 

discontinuities, defects and cracks at the root of the weld that would be rejectable 

under the AWS D1.1 provisions. W1a and W1b were distinguished from each 

other only by the size of the condition. Neither condition could be detected by 

visual inspection unless weld backing was removed, which, in the case of W1a 

conditions, would also result in removal of the original flaw or defect. At the time 

FEMA-267 was published, there was considerable controversy as to whether or 

not the various types of W1 conditions were actually damage or just previously 

undetected flaws introduced during the original construction. Research 

conducted since publication of FEMA-267 strongly supports the position that 

most, if not all W1 conditions are pre-existing defects, rather than earthquake 

damage. This research also demonstrated that W1 conditions are difficult to 

detect reliably unless the weld backing is removed. In a number of case studies, it 

has been demonstrated that when W1 conditions are indicated by UT, they are 

often found not to exist when weld backing is removed. Similarly, in other cases, 

upon removal of backing, W1 conditions were found to exist where none had been 

detected by UT. For these reasons, in the development of these recommendations, 

it has been decided to de-classify W1 conditions as damage and to eliminate the 

need for routine use of UT in the performance of detailed connection inspections. 

Notwithstanding the above, it is important to recognize that a very significant 

amount of the “damage” reported following the Northridge earthquake was type 

W1 conditions. Studies of 209 buildings in the city of Los Angeles have shown 

that approximately 2/3 of all reported “damage” conditions were type W1. 

Although these Recommended Criteria do not classify W1 conditions as damage, 

their presence in a connection can lead to a significant increase in the 

vulnerability of the building to earthquake induced connection fracture. If, in the 

performance of connection inspections or repairs it is determined that rejectable 

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Chapter 2: Inspection and Classification of Damage Steel Moment-Frame Buildings 

discontinuities, lack of fusion, slag inclusions or cracks exist at the root of a weld, 

they should be reported and consideration should be given to their repair, as a 

correction of an undesirable, pre-existing condition. 

Type W2 fractures extend completely through the thickness of the weld metal 

and can be detected by either magnetic particle testing (MT) or visual inspection 

(VI) techniques. Type W3 and W4 fractures occur at the zone of fusion between 

the weld filler metal and base material of the girder and column flanges, 

respectively. All three types of damage result in a loss of tensile capacity of the 

girder flange to column flange joint and should be repaired. 

2.2.4 Shear Tab Damage 

Six types of damage to girder-web-to-column-flange shear tabs are defined in Table 2-4 and 

illustrated in Figure 2-5. Severe damage to shear tabs is unlikely to occur unless other damage 

has also occurred to the connection, i.e., column, girder, panel zone, or weld damage, as 

previously defined. 

Table 2-4 Types of Shear Tab Damage 

Type Description 

S1 Partial crack at weld to column 

S2 Fracture of supplemental weld 

S3 Fracture through tab at bolts or severe distortion 

S4 Yielding or buckling of tab 

S5 Loose, damaged or missing bolts 

S6 Full length fracture of weld to column 

S1 

S3 S2

S4 

S6 

S5 

S1 

S3 S2

S4 

S6 

S5 

Figure 2-5 Types of Shear Tab Damage 

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Commentary: Shear tab damage should always be considered significant, as 

failure of a shear tab connection can lead to loss of gravity-load-carrying 

capacity for the girder, and potentially partial collapse of the supported floor. 

Severe shear tab damage typically does not occur unless other significant damage 

has occurred at the connection. If the girder flange joints and adjacent base 

metal are sound, they prevent significant differential rotations from occurring 

between the column and girder. This protects the shear tab from damage, unless 

excessively large shear demands are experienced. If these excessive shear 

demands do occur, then failure of the shear tab is likely to trigger distress in the 

welded joints of the girder flanges. 

2.2.5 Panel Zone Damage 

Nine types of damage to the column web panel zone and adjacent elements are defined in 

Table 2-5 and illustrated in Figure 2-6. This class of damage can be among the most difficult to 

detect since elements of the panel zone may be obscured by beams framing into the weak axis of 

the column. In addition, the difficult access to the column panel zone and the difficulty of 

removing sections of the column for repair, without jeopardizing gravity load support, make this 

damage among the most costly to repair. 

Table 2-5 Types of Panel Zone Damage 

Type Description 

P1 Fracture, buckle or yield of continuity plate 

P2 Fracture in continuity plate welds 

P3 Yielding or ductile deformation of web 

P4 Fracture of doubler plate welds 

P5 Partial depth fracture in doubler plate 

P6 Partial depth fracture in web 

P7 Full or near full depth fracture in web or doubler 

P8 Web buckling 

P9 Severed column 

P4 

P9 P2 P8 

P1 

P7 P3 P5, P6 

Figure 2-6 Types of Panel Zone Damage 

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Commentary: Fractures in the welds of continuity plates to columns (type P2), or 

damage consisting of fracturing, yielding, or buckling of the continuity plates 

themselves (type P1) may be of relatively little consequence to the structure, so 

long as the fracture does not extend into the column material itself. Fracture of 

doubler plate welds (type P4) is more significant in that this results in a loss of 

effectiveness of the doubler plate and the fractures may propagate into the column 

material. 

Although shear yielding of the panel zone (type P3) is not by itself 

undesirable, under large deformations such shear yielding can result in kinking of 

the column flanges and can induce large secondary stresses in the girder-flangeto-column-flange connection. 

Fractures extending into the column web panel zone (types P5, P6 and P7) 

have the potential, under additional loading, to grow and become type P9 (a 

complete disconnection of the upper half) of the column within the panel zone 

from the lower half, and are therefore potentially as severe as column splice 

failures. When such damage has occurred, the column has lost all tensile 

capacity and its ability to transfer shear is severely limited. Such damage results 

in a total loss of reliable seismic capacity. 

Panel zone web buckling (type P8) may result in rapid loss of shear stiffness 

of the panel zone with potential total loss of reliable seismic capacity. Such 

buckling is unlikely to occur in connections that are stiffened by the presence of a 

vertical shear tab for support of a beam framing into the column's minor axis. 

2.2.6 Other Damage 

In addition to the types of damage discussed in the previous sections, other types of structural 

damage may also be found in steel moment-frame buildings. Other framing elements that may 

experience damage include: (1) column base plates, beams, columns, and their connections that 

were not considered in the original design to participate in lateral force resistance, and (2) floor 

and roof diaphragms. In addition, large permanent interstory drifts may develop in structures. 

Based on observations of structures affected by the 1994 Northridge earthquake, such damage is 

unlikely unless extensive damage has also occurred to the lateral-force-resisting system. When 

such damage is discovered in a building, it should be reported and repaired, as suggested by later 

sections of these Recommended Criteria. 

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Steel Moment-Frame Buildings Chapter 3: Preliminary Postearthquake Assessment 

3. PRELIMINARY POSTEARTHQUAKE ASSESSMENT 

3.1 Introduction 

3.1.1 General 

Following a potentially damaging earthquake, an assessment should be performed for each 

steel moment-frame building to determine the likelihood of significant structural damage, the 

implications of this damage with regard to building safety and occupancy and the need for repair. 

A three-step process is recommended. These steps include: 

Screening. In this step, an estimate is made of the probable ground motion experienced at the 

building site. If this estimated ground motion falls below certain trigger values, further 

evaluation is not required. Section 3.2 provides recommended criteria for screening. 

Preliminary Evaluation. In this step, a site visit is made to the building and the condition of the 

building is observed to determine if there are obvious indications of structural or nonstructural 

damage that pose a potential risk to life safety. The building is typically posted with a placard, 

based on the findings of this evaluation. Section 3.3 provides recommended criteria for 

preliminary evaluation. 

Detailed Evaluation. In this step, detailed inspections of building framing and connections are 

performed to determine the condition of the structure. If structural damage is detected in the 

course of these inspections, further evaluations are performed to determine the significance of 

this damage and the appropriate repair and occupancy actions. Revision of the posting status of 

the building may be appropriate following such evaluation. Chapters 4 and 5 provide procedures 

for detailed evaluation. 

As indicated in Section 1.3 and Chapter 2, following the 1994 Northridge and other recent 

earthquakes, structural damage was detected in many steel moment-frame buildings that had little 

outward signs of structural distress. Detailed postearthquake evaluations are necessary to find 

such damage, but involve rigorous inspection of structural condition. These more detailed 

evaluations can be quite costly and may be unnecessary for buildings that have not sustained 

significant structural damage. The initial screening process presented in this chapter is intended 

to provide rapid identification of those buildings that likely did not experience sufficient ground 

shaking to cause significant damage and which therefore need not be subjected to further 

evaluations. The preliminary evaluation procedures of this chapter are intended to identify those 

buildings that present obvious signs of severe damage so that immediate restrictions on 

occupancy may be placed. Following preliminary evaluation, a report of pertinent findings 

should be made to the Owner. If the evaluation was ordered by the Building Official, these 

findings should also be reported to the Building Department and the building should be posted 

with an appropriate placard in accordance with Section 3.3. 

Commentary: Screening is intended to identify those buildings that experienced 

sufficient ground shaking that they may have sustained significant damage. If a 

building is not identified as likely to have experienced such ground shaking, no 

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evaluation need be performed. However, if a building is identified as having 

experienced such ground motion both a preliminary and detailed evaluation 

should be performed. The preliminary evaluation is intended to provide a rapid 

basis for making recommendations regarding immediate postearthquake 

occupancy. Detailed evaluations, in accordance with Chapters 4 and 5 are used 

to confirm the extent and severity of any damage present and to serve as the basis 

for repair programs, should these be necessary. 

The procedures contained in this chapter and in Chapters 4 and 5 are specifically 

intended to identify if earthquake ground shaking has damaged a building and 

thereby, impaired its safety. These procedures are not intended to determine if a 

building had adequate structural characteristics prior to the onset of damage or 

how the structure may perform in future earthquakes. Some owners may wish to 

assess the likely performance of their building when subjected to a future 

earthquake, irrespective of any damage that has occurred in the present event. 

Readers are referred to the companion publication, FEMA-351 – Recommended 

Seismic Evaluation and Upgrade Criteria for Existing Welded Steel MomentFrame Buildings for performance evaluation and upgrade recommendations for 

such structures. It is recommended that such performance evaluations be 

performed when a building has sustained substantial damage as a result of 

ground shaking that is significantly less intense than the shaking specified by the 

current building code for design of a new structure at that site. 

3.1.2 Evaluator Qualifications 

Postearthquake evaluations entail the observation of different conditions within a building, 

making judgments as to whether they are indicative of structural damage and the likely effect of 

such damage with regard to the ability of the structure to withstand additional loading. This 

requires the application of considerable structural engineering knowledge and judgment. In order 

to perform these tasks properly, the evaluator should possess at least the same levels of 

knowledge, experience and training necessary to act as the design professional of record for the 

structure, and in some cases, more detailed knowledge, experience and training may be 

necessary. Persons possessing such knowledge, experience and training are referred to in these 

Recommended Criteria as the structural engineer. References to the structural engineer 

throughout these Recommended Criteria indicate that the work is to be performed either directly 

by persons possessing these qualifications, or by persons acting under the direct supervision of 

such a person. 

3.2 Screening 

Prior to performing preliminary or detailed postearthquake evaluations, it is recommended 

that screening be performed to determine if a building has likely experienced ground shaking of 

sufficient intensity to cause significant damage. Buildings need to be subjected to evaluations 

only if any of the following apply: 

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• estimated ground-motion acceleration or intensity (MMI) at the site exceeds the limits 

indicated in Table 3-1; 

• significant structural damage is observed in one or more steel moment-frame structures 

located within 1 kilometer of the building on sites with similar or more firm soil profiles; 

• significant structural damage is observed to one or more modern, apparently well-designed 

structures (of any structural system) within 1 kilometer of the building and on sites with 

similar or more firm soil profiles; 

• damage to the general building stock within 1 kilometer of the building and on sites with 

similar or more firm soil profiles corresponds with the categories indicated in Table 3-1; 

• for an earthquake having a magnitude of 6.5 or greater, the structure is either within 5 

kilometers of the trace of a surface rupture or within 5 kilometers of the ruptured area of the 

fault plane when no surface rupture has occurred; 

• significant architectural or structural damage is observed in the building; or 

• entry to the building has been limited by the building official because of earthquake damage, 

regardless of the type or nature of the damage. 

Table 3-1 Ground Motion Indicators of Potential Damage 

1997 NEHRP MCE 

Map* 

Short-Period Contour 

Area 

Estimated Peak 

Ground 

Acceleration 

Level of Damage to Buildings Within 1 

Kilometer 

Estimated 

Modified Mercali 

Intensity, MMI 

SS > 0.50 > 0.25g Prevalent partial collapse of unreinforced 

masonry buildings. High levels of 

nonstructural damage. Considerable 

damage to ordinary buildings. 

VIII 

0 < SS < 0.50 > 0.15g Considerable damage to unreinforced 

masonry buildings. Slight damage to welldesigned buildings. 

Prevalent nonstructural damage. 

VII 

* FEMA-302, ASCE (1998) and IBC (ICC, 2000) maps. 

If none of the above conditions apply to a building, it may be classed as unlikely to have 

experienced significant damage and need not be subjected to evaluation. 

Commentary: Preliminary screening is typically performed by the Building 

Official in order to identify those areas of a community in which post earthquake 

evaluations should be performed. The screening criteria presented in this section 

can typically be applied on a regional basis, after preliminary reconnaissance has 

been performed to determine the general patterns and distribution of damage that 

has occurred in the affected region. Building departments will typically perform 

such surveys in the hours immediately following an earthquake, in coordination 

with emergency response agencies in order to coordinate emergency response 

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activities. The data obtained from such surveys can be used to develop 

preliminary isoseismal maps (maps with contours indicating probable intensities 

of ground shaking). These isoseismal maps can then be used to identify 

geographic areas, within which evaluations should be performed or ordered. 

Typically, initial determination of the distribution of ground motion intensity from 

an earthquake and the geographic areas in which building evaluations should be 

performed will be subject to revision, over time, as more detailed data becomes 

available. A number of techniques and sources of information are available for 

developing these more accurate estimates of ground motion intensity. Frequently, 

the United States Geologic Survey (USGS) or other government agencies will 

develop maps of ground motion intensity, shortly after an earthquake occurs. In 

regions with a large number of strong-motion accelerographs present, actual 

ground motion recordings provide the best method of mapping contours of 

ground motion. These should be used if located near the building, and are 

located on sites having similar characteristics. 

In other regions, empirical techniques, such as the use of standard ground-motion 

attenuation relationships (e.g., Boore and Joyner, 1994; Campbell and 

Bozorgnia, 1994) may be required. These can be supplemented with analytically 

derived estimates such as those obtained by direct simulation of the fault rupture 

and ground wave propagation. It may be desirable to retain a qualified 

geotechnical engineer or earth science consultant to make these estimates. It 

should be noted, however, that lacking direct instrumental evidence, site-specific 

ground motion estimates are, at best, uncertain and subject to wide variations 

depending on the assumptions made. Therefore, the best indicator of the severity 

of ground motion at a site is often the performance of adjacent construction. The 

criteria of Table 3-1 are provided to help assure that sites that experienced 

relatively strong ground motion are not overlooked as a result of inaccurate 

estimates of the ground motion severity. 

3.3 Preliminary Evaluation 

3.3.1 General 

The objective of preliminary evaluation is to determine, on a rapid, preliminary basis, 

whether a building has sustained either structural or nonstructural damage that results in a 

hazardous condition. Preliminary evaluation includes: 

• a general review of the building’s construction characteristics to determine its structural 

system and vulnerable features (Section 3.3.2), 

• a visit to the building site to observe its overall condition and note obvious signs of damage 

(Section 3.3.3), 

• a determination of an appropriate posting category for the building, on the basis of the 

preceding results and engineering judgment (Section 3.3.4). 

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The condition ratings presented in Table 3-2 are recommended as posting categories. Section 

3.3.4 provides recommended criteria for assignment of a building to the various posting 

categories. 

Table 3-2 Postearthquake Condition Designations 

Condition Finding Description 

G 

R 

E 

E 

N 

1 Inspected The building does not appear to have experienced significant damage either 

to structural or nonstructural components. Occupancy may continue, pending 

completion of detailed evaluations. 

2 Minor nonstructural 

damage 

The building does not appear to have experienced significant damage to 

structural elements, but has experienced some damage to nonstructural 

components. Occupancy may continue, pending completion of detailed 

evaluations. Repair of nonstructural damage may be conducted at 

convenience. 

3 Minor damage The building appears to have sustained limited damage to structural and 

nonstructural elements. Occupancy may continue, pending completion of 

detailed evaluations. Repair of damage may be conducted at convenience. 

Y 

E 

L 

L 

O 

W 

1 Damaged – 

nonstructural 

The building does not appear to have experienced significant damage to 

structural elements; however, it has sustained damage to nonstructural 

components that pose a limited safety hazard. Occupancy of the building in 

areas subject to these hazards should be limited until repairs are instituted. 

Occupancy of other portions of the building may continue, pending 

completion of detailed evaluations. 

2 Damaged – 

structural 

The building appears to have experienced significant damage to structural 

elements. Although it does not appear that the building is an imminent 

collapse risk, localized safety hazards may exist. Occupancy of the building 

in areas subject to these hazards should be limited until repairs or 

stabilization can be implemented, or a more reliable assessment of the 

building’s condition can be made to demonstrate that hazards do not exist. 

Occupancy of other portions of the building may continue, pending 

completion of detailed evaluations. 

R 

E 

D 

1 Unsafe – repairable The building appears to have sustained significant damage to structural 

elements that has substantially impaired its ability to resist additional loading 

or to nonstructural elements that pose a significant hazard to occupants. It 

should not be occupied until repair or stabilization work has been performed 

or a more detailed evaluation of its condition can be made to demonstrate 

that hazards do not exist. 

2 Unsafe The building appears to have sustained significant damage to structural 

elements, substantially impairing its ability to resist additional loading. It 

appears to be a potential collapse hazard and should not be occupied. 

Commentary: The condition assessment categories indicated in Table 3-1 should 

be assigned on the basis of the preliminary(rapid) evaluation. However, the 

assignment should be subject to change on the basis of detailed evaluations 

conducted in accordance with Chapters 4 and 5. 

It is not uncommon during the postearthquake evaluation process to discover 

that although a building has relatively little damage, it has severe structural 

deficiencies relative to current building code requirements and may as a result be 

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structurally unsafe. The condition assessments indicated in Table 3-2 are 

intended to be applied only to those conditions resulting from earthquake damage 

and should not be used to rate a building that is otherwise structurally deficient. 

However, when such deficiencies are identified in a building during the course of 

a postearthquake evaluation, the engineer should notify the Owner and Building 

Official of these conditions. 

3.3.2 Building Construction Characteristics 

In order to make a meaningful assessment of a building's postearthquake condition it is 

necessary to develop an understanding of its structural system and basic details of the building’s 

construction and to clearly establish the seismic load path. Whenever the structural and 

architectural drawings for the building are available, they should be reviewed as part of the 

preliminary evaluation. The review should include the following: 

• confirmation that the building is a steel moment-frame structure, 

• determination of the year of design and construction and code used as a basis; this may 

provide information on particular vulnerabilities, such as the presence of weak stories, or use 

of particular weld metals, 

• identification of materials and typical details of connections and elements for areas of 

particular vulnerability, 

• identification of the location of steel moment frames, 

• identification of locations of moment-resisting beam-column connections and column 

splices, to identify locations where potentially vulnerable conditions exist, 

• identification of any structural irregularities in the vertical and horizontal load resisting 

systems, that could lead to potential concentrations of damage, and 

• identification of architectural elements that could affect the behavior of the structural system 

or elements, or that may themselves be vulnerable to damage and be a threat to occupants, 

including, for example, precast concrete cladding systems and interior shaft walls. 

3.3.3 Preliminary Site Inspection 

Every steel moment-frame building situated on a site that has experienced strong ground 

shaking, as identified in accordance with the screening criteria of Section 3.2, should be 

subjected to a rapid postearthquake inspection to ascertain whether there is apparent damage and 

to determine the apparent severity of such damage. When performing the inspection, the 

structural engineer should attempt to determine if strong motion accelerometers are present in the 

building. If so, the record should be accessed and reviewed for noticeable changes in behavior 

during the building response that may be indicative of significant structural damage. 

Preliminary site inspections should include the following: 

1. Visual observation of the building exterior. Check for: 

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� obvious indications of large permanent interstory drift, 

� indications of foundation settlement or distress as evidenced by sags in horizontal 

building fenestration or distress in base level slabs, 

� loosened or damaged cladding or glazing systems, 

� indications of discrete areas of the building where interstory drift demands may have 

concentrated as evidenced by apparent concentrations of architectural damage to fascia 

and cladding systems, 

� pounding against adjacent buildings or portions of the building separated by expansion 

joints, and 

� potential site instabilities such as landslides or lateral spreading that may have resulted in 

damage to the building foundations or structure. 

2. Visual observation of the building interior. Check for: 

� damage to nonstructural components, such as suspended ceilings, light fixtures, ducting, 

and masonry partitions, that could result in potential hazards, 

� damage to floor slabs around columns, to finishes, and to partitions, that may suggest 

damage to adjacent beams and connections, 

� indications of discrete areas of the building where interstory drift demands may have 

concentrated as evidenced by apparent concentrations of damage to architectural elements 

including interior partitions, 

� damage to interior finishes on structural elements, such as columns, that could be 

indicative of damage to the underlying structure, 

� damage to equipment or containers containing potentially hazardous substances, and 

� damage to elevator counterweight and rail systems. 

3. Evaluation of the building for permanent interstory drift. 

Preliminary evaluation of the building for permanent interstory drift should be performed. 

This can be done by dropping a plumb bob through the elevator shaft and determining any 

offset between threshold plates in adjoining levels of the building. Multiple levels should be 

checked simultaneously, to minimize the effect of minor offsets resulting from withintolerance variations in the original construction. 

4. Perform preliminary visual inspection of selected moment frames for indications of damage. 

Refer to Sections 3.3.3.1 and 3.3.3.2 for preliminary inspection procedures for momentresisting connections with and without fireproofing present, respectively. 

� If visual observation of building exterior or interior indicates a zone or zones of large 

permanent interstory drift, perform selective removal of architectural finishes to expose 

framing. Observe for indications of yielding, buckling or other damage to framing, or 

connections. Exposures and observation should be made of at least one beam-column 

connection per line of framing per story within the zone or zones of large permanent 

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interstory drift. For highly redundant structures, with many lines of framing, exposures 

and observations may be limited to one beam-column connection on one line of framing 

in each direction of building response, on each side of the structure, with a minimum of 

two exposures per story. 

� If visual observation of building exterior or interior indicates zones of concentrated 

interstory drift demand, perform selective removal of architectural finishes to expose 

framing. Observe for indications of fracture, yielding or buckling of framing, or damage 

to connections. Exposures and observation should be made of at least one beam-column 

connection per line of framing per story within the zone or zones of concentrated 

interstory drift demand. For highly redundant structures, with many lines of framing, 

exposures and observations may be limited to one beam-column connection on one line 

of framing in each direction of building response, on each side of the structure, with a 

minimum of four exposures per story. 

� If visual observation of building exterior or interior indicates neither zones of large 

permanent interstory drift, nor of concentrated interstory drift demand, perform selective 

removal of architectural finishes to expose framing throughout structure. Observe for 

indications of yielding or buckling of framing, or damage to connections. Exposures and 

observation should be made of at least one beam-column connection per line of framing 

per story. For highly redundant structures, with many lines of framing per story, 

exposures and observations may be limited to one beam-column connection on one line 

of framing in each direction of building response, on each side of the structure, with a 

minimum of two exposures per story. 

� If visual observation of the building exterior indicates zones of pounding against adjacent 

structures, expose framing in the area of pounding to identify damage to structural 

elements and connections. 

Commentary: In most steel moment-frame buildings, structural steel will be 

obscured by fire protective coverings that are frequently difficult to remove. In 

many cases these coverings will be composed of asbestos-containing materials, if 

constructed before 1976, and must not be removed by anyone without proper 

training. Observation conducted as part of preliminary procedures is limited to 

observing the condition of the steel, if exposed to view, or the condition of the fire 

protective covering if the steel is not exposed, to observe tell-tale signs of 

structural damage including cracking or spalling of the covering material, or 

loosened and broken bolts. 

The presence of one or more strong-motion instruments in a building can 

provide valuable evidence as to the extent of damage a building has experienced. 

Noticeable lengthening of the building period can be an indication of structural 

damage. However, even in the absence of instruments within a building, it may be 

possible to obtain indirect evidence of changes in a building’s dynamic properties 

that are indicative of damage. This could include apparent lengthening of the 

building period, or increasing nonstructural damage in aftershocks. 

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3.3.3.1 Preliminary Connection Inspections when Fireproofing is Present 

Perform the observations indicated in the checklist below. Figure 3-1 indicates the various 

zones of observation. Note that fireproofing need not be removed as part of the preliminary 

inspection, unless indications of potential damage are noted, at which point fireproofing should 

be removed to allow confirmation of the extent of any damage. If there is reason to believe the 

fireproofing is an asbestos-containing material, removal should be performed by appropriately 

trained personnel with proper personnel protection. The engineer should not personally attempt 

to remove fireproofing suspected of being an asbestos containing material unless he has been 

trained in the appropriate hazardous materials handling procedures and is wearing appropriate 

protective equipment. 

Beam top flange 

Beam bottom flange 

Joint of beam flange to column 

Shear tab region 

Panel Zone 

Continuity plates 

Figure 3-1 Observation Zones for Fire-Proofed Beam-Column Connections 

� Observe beam framing into connection for trueness to line, and potential indications of lateral 

flexural-torsion buckling (damage type G8, Section 2.2.1). 

� Observe condition of fireproofing along beam within one beam depth of the column for 

cracking or spalling of the fireproofing material along the beam surface, indicating potential 

yielding or buckling of the beam flanges (damage types G1 and G2, Section 2.2.2). 

� Observe the top and bottom surface of the bottom flange fireproofing and bottom surface of 

the top flange fireproofing at the locations where the beam flanges join the column flanges 

(or continuity plates for minor axis connections) for cracks or losses of material that could 

indicate cracking at the full penetration weld (damage types G3, Section 2.2.1; C1, C3 and 

C4, Section 2.2.2; W2, W3, W4, Section 2.2.3). 

� Observe the condition of the fireproofing at the beam web, in the vicinity of the clip 

connection from the beam web to the column for loosened, cracked or spalled material 

indicative of potential damage to shear tabs (damage types S1 through S5, Section 2.2.4). 

� Observe the condition of the fireproofing at the column panel zone for cracks, loosened or 

spalled material, indicative of damage to the panel zone or continuity plates (damage types 

P1 through P8, Section 2.2.5). 

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� Observe the flanges of the column at and beneath the joint with the beam flange for loosened, 

spalled or cracked material, indicative of fractures, buckled or yielded sections (damage types 

C1, C3, C4, C6, Section 2.2.6). 

� Observe the column flange in the area immediately above the bottom beam flange for 

loosened, spalled or cracked material, indicative of a potential divot type fracture of the 

column material (damage type C2, Section 2.2.2). 

Commentary: The presence of fireproofing will tend to obscure many types of 

damage, unless the damage is very severe. However, removal of fireproofing can 

be a difficult and time consuming process. For the purposes of preliminary 

inspection in buildings with fireproofing, inspection is limited to that readily 

observable with the fire proofing in place. Removal of fireproofing and more 

careful visual inspection in such buildings is limited to inspections performed as 

part of detailed evaluations, in accordance with Chapters 4 and 5 of this 

publication. An exception is the case when observation indicates that the 

fireproofing has noticeably cracked, spalled or loosened, indicating that damage 

has probably occurred to the steel framing beneath. In this case, removal of 

fireproofing is recommended as part of the preliminary inspection to determine 

the extent of damage. 

In many buildings constructed prior to 1976, the original fireproofing materials 

commonly contained friable asbestos fibers. Disturbing such material without 

wearing suitable breathing apparatus can result in a significant health hazard 

both to the person performing the work and also to others located in the area. 

For this reason, owners have been gradually addressing these hazards either by 

encapsulating such fireproofing, to prevent it from being disturbed, or replacing 

it with non-hazardous materials. In buildings constructed prior to 1976, the 

engineer should not permit fireproofing to be removed except by properly trained 

personnel using appropriate procedures unless the owner can present suitable 

evidence that the material does not contain friable asbestos. 

3.3.3.2 Bare Structural Steel 

Preliminary inspection of framing connections in buildings that do not have fireproofing in 

place on the structural steel should include the complete joint penetration (CJP) groove welds 

connecting both top and bottom beam flanges to the column flange, the backing bars and the 

weld access holes in the beam web; the shear tab connection, including the bolts, supplemental 

welds and beam web; the column web panel zone, including doubler plates; and continuity plates 

and continuity plate welds (see Figure 3-2). 

The inspection should be by visible means. Observe all exposed surfaces for cracks, 

buckling, yielding, and loosened or broken bolts. The area inspected should include that portion 

of the beam within a distance db (beam depth) of the face of the column, that portion of the 

column below the connection and within a distance dc (column depth) of the bottom beam flange, 

the panel zone and all bolts and plates within these regions. Sections 2.2.1 through 2.2.6 indicate 

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the types of damage that may be present. All damage observed should be recorded according to 

the classification indicated in those sections, and documented in sketch form. 

Zone 

Beam 

Column 

Backing 

Shear Tab 

Supplemental weld 

Panel 

Continuity Plates 

Figure 3-2 Components of Moment Connection 

Note that visual inspection should not be performed casually. After a fracture forms in steel 

framing, it can close up again under further loading of the building. Such “closed” fractures, 

though obscure, can typically be detected by careful observation, sometimes aided with touch to 

detect roughness in the surface in the vicinity of a potential fracture. Wetting of the area of a 

suspected surface crack can also assist in detection. In some cases, it may be necessary to use 

more formal nondestructive testing methods, such as ultrasonic testing, magnetic particle testing, 

or liquid dye penetrant testing to confirm the presence of such cracks. Such confirmation can be 

performed as part of the more detailed inspections undertaken as part of a Level 1 detailed 

evaluation (Chapter 4) or a Level 2 detailed evaluation (Chapter 5). 

Certain types of damage (C2, C3, C5, Section 2.2.2; W2, W3, Section 2.2.3) may be 

impossible to detect by visual observation alone, as the presence of weld backing at the underside 

of the beam flange will obscure the presence of the fracture. The presence of a gap between the 

bottom edge of the backing and the column flange is one indication of the potential presence of 

such damage. If such a gap is present it may be possible to explore the presence of concealed 

fractures by inserting a feeler gauge into the gap to determine its depth. If the feeler gauge can be 

inserted to a depth that exceeds the weld backing thickness, a fracture should be assumed to be 

present. Nondestructive testing will be required to confirm the extent of such damage, and can 

be performed as part of the more detailed evaluation. Alternatively, the backing can be removed 

to allow direct observation of any damage present. However, such removal entails either cutting 

or grinding operations and can not normally be performed as part of a preliminary evaluation. 

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3.3.4 Data Reduction and Assessment 

Following the collection of data on a building, as outlined in Sections 3.3.1 and 3.3.2, it is 

necessary to form a preliminary opinion as to whether a building has sustained damage that 

creates a potential hazard, and the severity and distribution of such hazards, if present. The 

following sections provide recommendations in this regard. The structural engineer, on the basis 

of the evaluated data, or personal engineering judgment, may make a more conservative 

assessment. 

3.3.4.1 Finding of Dangerous Condition 

An assessment should be made that a building has been extensively damaged and is 

potentially hazardous, if any of the following conditions are observed: 

• permanent interstory drift in any level of 1.0% or greater, 

• unexpected severe damage to architectural elements or significant period lengthening of the 

building is observed in aftershocks, 

• visual inspections of steel framing indicate the presence of two or more fractures of the type 

G7, C3, C6, C7, S3, S4, S5, S6, P6, P7 or P9, at any floor level, or 

• the building experiences excessive lateral deformation or unusual amounts of additional 

architectural damage in moderate aftershocks. 

In the event that any of the above conditions is detected, the building should be assessed on a 

preliminary basis as conforming to damage condition Red-1, of Table 3-2. A detailed evaluation 

should be recommended and notification should be made advising against continued occupancy 

until a more detailed determination of structural condition can be completed. 

Commentary: The observed behavior of a building in repeated aftershocks may 

provide some clues as to whether it has experienced significant structural 

damage. In instrumented buildings it may be possible to observe a lengthening of 

the building period during aftershocks. In buildings without instruments, the 

observation of unexpected large amounts of architectural damage during 

aftershocks could indicate the presence of previous structural damage. 

3.3.4.2 Finding of Damaged Condition 

If none of the conditions indicated in Section 3.3.4.1 are determined to exist, but one or more 

of the conditions indicated below are present, an assessment should be made that the building has 

sustained significant nonstructural damage and should be posted as damage condition Yellow-1 

of Table 3-2. Appropriate precautions should be taken to limit access to hazardous areas. 

• Connections of exterior fascia panels have been damaged and panels are hanging loosely on 

the building. 

• Exterior glazing is broken above the first story. 

• Connections of stair stringers to floor framing has been compromised. 

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• Ceiling components, including suspension systems, lights, HVAC and utilities have been 

damaged and are hanging into the occupied spaces or walkways. 

• Gas lines are damaged or containers of unidentified or known hazardous materials have 

toppled and spilled. 

• Egress ways are blocked or inoperable. 

• Emergency lighting systems are unusable. 

• Fire suppression systems required by code are inoperable. 

If none of the conditions indicated in Section 3.3.4.1 are determined to exist, but one or more 

of the conditions indicated below are present, an assessment should be made that the building has 

sustained significant structural damage and should be posted as damage condition Yellow-2 of 

Table 3-2. Appropriate precautions should be taken to limit access to hazardous areas. 

• Visual inspection of steel framing indicates shear tab damage type S3, S5 or S6 in any beam 

connection, in accordance with Section 2.2.4. 

• Visual inspection of steel framing indicates that a beam has become dislodged from a 

supporting member or element. 

• Visual inspection of steel framing indicates that a column has experienced type P7 damage in 

accordance with Section 2.2.5, or type C7 damage in accordance with Section 2.2.2. 

3.3.4.3 Finding of Undamaged Condition 

If none of the conditions indicated in Sections 3.3.4.1 or 3.3.4.2 are determined to exist, it is 

recommended that the building be assessed Green-1, Green-2, or Green-3 of Table 3-2, as 

appropriate, pending completion of detailed evaluations in accordance with Chapter 4 or 5. 

Commentary: The absence of significant observable damage to steel momentframe structures in a preliminary evaluation on sites believed to have experienced 

strong ground motion, per Table 3-1, should not be used as an indication that 

detailed evaluations are not required. Many steel moment-frame buildings that 

were structurally damaged by the 1994 Northridge and 1989 Loma Prieta 

earthquakes had little apparent damage based on casual observation. 

3.3.5 Reporting and Notification 

Following performance of a preliminary evaluation, notification should be made that an 

evaluation has been performed and a report should be provided to the Owner. The extent of 

notification to be made is dependent upon the jurisdiction of the party performing the evaluation, 

and upon the condition of the building. If the building has been found to be dangerous, the 

occupants ultimately must be notified (in a timely manner). 

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3.3.5.1 Building Departments 

When preliminary evaluations are performed by or on behalf of the Building Official, or other 

authority having jurisdiction, the following notifications should be made: 

• A placard should be placed at the main entry to the building indicating that a preliminary 

(rapid) evaluation has been performed, and indicating the assessed condition designation of 

the building, recommended occupancy restrictions, follow-up actions, and the identity and 

affiliation of the person performing the evaluation. In large buildings with more than one 

entrance, additional placards should be placed at all other entrances (ATC, 1989). Appendix 

B to these Recommended Criteria includes sample placards (from ATC, 1995). 

• If a building has been posted either as "damaged" (condition Yellow-1 or Yellow-2) or 

"unsafe" (condition Red-1 or Red-2), additional written notification should be served on the 

Owner at his/her legal address, indicating the status of the posting, the Owner's rights and any 

actions required on the Owner's part. 

3.3.5.2 Private Consultants 

If postearthquake evaluations by private consultants are permitted by the local authority 

having jurisdiction, the same procedures prescribed in Section 3.3.5.1 should be followed: a 

placard should be placed at the main entry to the building indicating that preliminary evaluation 

has been performed, the assessed condition of the building, recommended occupancy restrictions 

and follow-up actions, and the identity and affiliation of the person performing the evaluation. In 

large buildings with more than one entrance, additional placards should be placed at all other 

entrances (ATC, 1989). Appendix B to these Recommended Criteria includes sample placards 

(from ATC, 1995). 

In addition, a formal report should be prepared indicating the scope of evaluation that has 

been performed, the findings of the evaluation, including a description of any damage 

encountered, the appropriate postearthquake condition designation assigned to the building and 

any recommendations for additional evaluation, restrictions of occupancy and/or repair action. 

The report should be submitted to the party requesting the evaluation and to other parties as 

required by law. 

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4. LEVEL 1 DETAILED POSTEARTHQUAKE EVALUATIONS 

4.1 Introduction 

Detailed evaluation is the second step of the postearthquake evaluation process. It should be 

performed for all buildings that are estimated to have experienced potentially damaging ground 

motions, using the screening procedures of Section 3.2 of these Recommended Criteria. As 

detailed evaluation can be a time consuming process, it is recommended that a preliminary 

evaluation, in accordance with the procedures of Chapter 3, be conducted prior to detailed 

evaluation, to permit rapid identification of those buildings that may have been so severely 

damaged that they pose an immediate threat to life safety. 

Many steel moment-frame buildings damaged in past earthquakes have displayed few 

outward signs of structural or nonstructural damage. Consequently, except for those structures 

which have been damaged so severely that they are obviously near collapse, brief evaluation 

procedures, such as those of Chapter 3, are unlikely to provide a good indication of the extent of 

damage or its consequences. In order to make such determination, it is necessary to perform 

detailed inspections of the condition of critical structural components and connections. If 

structural damage is found in the course of such inspections, it is then necessary to make a 

determination as to the effect of discovered damage on the structure’s ability to resist additional 

loading. Ultimately, decisions as to the significance of damage, whether occupancy should be 

permitted in a building and whether specific types of damage should be repaired must be made 

on the basis of quantitative evaluation and engineering judgment. 

This chapter provides simplified procedures for a quantitative evaluation method in which 

occupancy and repair decisions are assisted based on the calculation of a damage index, related to 

the distribution and severity of different types of damage in the structure. In order to apply this 

method, termed a Level 1 evaluation, it is necessary to obtain an understanding of the distribution 

of damage in the structure. This must be obtained by performing visual inspections of critical 

framing and connections. It is preferred that damage indices be calculated based on a 

determination of the condition of all critical connections in the building; however, it is 

permissible to infer a distribution of damage, and calculate a damage index, based on an 

appropriately selected sample of connections. 

Chapter 5 provides recommended criteria for an alternative method of quantitative 

evaluation, termed a Level 2 evaluation, based on performing structural analysis of the damaged 

structure’s ability to resist additional strong ground shaking. In order to perform a Level 2 

evaluation, it is necessary to conduct a complete inspection of all fracture-susceptible 

connections in the building. 

Commentary: The Level 1 evaluation approach contained in this chapter is based 

upon a methodology originally presented in FEMA-267, modified to account for 

experience gained in the application of the FEMA-267 guidelines to real 

buildings and also calibrated to expert opinion on the severity of various types of 

damage. The Level 2 evaluation is a more comprehensive approach that is 

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Chapter 4: Level 1 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

compatible with the overall approach developed for performance evaluation of 

structures. 

The Level 1 detailed evaluation procedure consists of gathering available 

information on construction of the structure and a multi-step inspection, 

evaluation, decision and reporting process. Although it is preferable to conduct a 

complete inspection of all fracture-susceptible connections, it is permissible to 

inspect only a selected portion of the elements and connections and to use 

statistical methods to estimate the overall condition of the building. A damage 

index is introduced to quantify the severity of damage in the building. This 

damage index is calculated based on individual connection damage indices, di, 

assigned to the individual inspected connections. These connection damage 

indices vary between 0 and 4, with 0 representing no significant earthquake 

damage and 4 representing severe damage. A story-level damage index, Dmax, is 

introduced which varies between 0 and 1.0, depending on the severity of damage. 

Based on the maximum damage index obtained for any floor level, Dmax, or if full 

inspections were not made of all connections, the probability that the damage 

index exceeds a specified threshold, recommendations are provided to the 

structural engineer regarding the appropriate damage condition designation as 

well as decisions regarding occupancy restrictions and repair actions. 

4.2 Data Collection 

Prior to performing a detailed inspection and evaluation, available information on the 

building’s construction should be collected and reviewed. This review should be conducted in a 

manner similar to that indicated in Section 3.3.2, but extended to include greater knowledge, for 

example, of the primary lateral and gravity load-resisting systems, typical detailing, and presence 

of irregularities. Pertinent available engineering and geotechnical reports, including any previous 

damage survey reports, such as the preliminary postearthquake evaluation report prepared in 

accordance with Chapter 3 of these Recommended Criteria, and current ground motion estimates, 

should also be reviewed. Specifications (including the original Welding Procedure 

Specifications), shop drawings, erection drawings, and construction inspection records should be 

reviewed when available. 

When structural framing information is not available, a comprehensive field study should be 

undertaken to determine the location and configuration of all lateral-force-resisting frames, and 

the details of their construction, including members’ sizes, material properties, and connection 

configurations. See Section 5.2 for additional discussion. 

4.3 Evaluation Approach 

Analyses of buildings with brittle connections, such as those damaged by the 1994 

Northridge earthquake, show that although damage occurs slightly more often in locations 

predicted by analysis to have high stress and deformation demands, damaged connections tend to 

be widely distributed throughout building frames, often at locations that analyses would not 

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predict. This suggests that there is some randomness in the distribution of the damage. To detect 

reliably all such damage, it is necessary to subject each fracture-susceptible connection to 

detailed inspections. Fracture-susceptible connections include: 

• Moment-resisting beam-column connections in which the beams are connected to columns 

using full penetration welds between the beam flanges and column, and in which yield 

behavior is dominated by the formation of a plastic hinge within the beam at the face of the 

column, or within the column panel zone. 

• Splices in exterior columns of moment-resisting frames when the splices consist of partial 

penetration groove welds between the upper and lower sections of the column, or of bolted 

connections that are incapable of developing the full strength of the upper column in tension. 

The inspection of all such connections within a building can be a costly and disruptive 

process. Although complete visual inspections of fracture-susceptible connections are 

recommended as part of a Level 1 evaluation, this evaluation methodology permits a 

representative sample of the critical connections to be selected and inspected. When only a 

sample of connections is inspected use is made of statistical techniques to project damage 

observed in the inspected sample to that likely experienced by the entire building. 

In order to obtain valid projections of a building’s condition, when the sampling approach is 

selected, samples should be broadly representative of the varying conditions (location, member 

sizes, structural demand) present throughout the building and samples should be sufficiently 

large to permit confidence in the projection of overall building damage. Two alternative methods 

for sample selection are provided. When substantial damage is found within the sample of 

connections, additional connections should be inspected to provide better, more reliable 

information on the building condition. 

Once the extent of building damage is determined, (or estimated if a sampling approach is 

utilized) the structural engineer should assess the residual structural capacity and safety, and 

determine appropriate repair and/or modification actions. General recommendations are 

provided, based on calculated damage indices. As an alternative to this approach, direct 

application of engineering analysis (Level 2 evaluation) may also be used as provided for in 

Chapter 5 of these Recommended Criteria. 

4.4 Detailed Procedure 

Postearthquake evaluation should be carried out under the direct supervision of a structural 

engineer. Two alternative procedures are presented below depending on whether all connections 

in the building are inspected, or only a sample of the connections in the building are inspected. 

Section 4.4.1 describes the procedure when all connections are inspected. Section 4.4.2 

describes the procedure when a sample of connections are inspected. 

As used in these Recommended Criteria, the term “connection” means that assembly of 

elements including the beam, column, plates, bolts, and welds, that connect a single beam to a 

single column. Interior columns of plane frames will typically have two connections (one for 

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each beam framing to the column) at each floor level. Exterior columns of plane frames will 

have only one connection at each floor level. 

4.4.1 Method 1 - Inspection of All Connections 

The following five-step procedure may be used to determine the condition of the structure 

and to develop occupancy, repair and modification strategies when all critical connections in a 

building are inspected and the extent of damage to all connections is known: 

Step 1: Conduct a complete visual inspection of all fracture-susceptible connections in the 

building in accordance with Section 4.3. Moment-resisting connections should be 

inspected in accordance with Section 4.4.1.1, with supplemental nondestructive 

examination, as suggested in that section. 

Step 2: Assign a connection damage index, di, to each inspected connection in accordance 

with Section 4.4.1.2. 

Step 3: Calculate the floor damage index at each floor, Dj, pertinent to lateral force resistance 

of the building in each of two orthogonal directions, in accordance with Section 

4.4.1.3. Determine the maximum of the floor damage indices, Dmax. 

Step 4: Based on the calculated floor damage indices, determine appropriate occupancy, and 

structural repair strategies, in accordance with Section 4.4.1.5. If deemed appropriate, 

the structural engineer may conduct detailed structural analyses of the building in the 

as-damaged state, to obtain improved understanding of its residual condition and to 

confirm that the recommended strategies are appropriate or to suggest alternative 

strategies. Recommendations for such detailed evaluations are contained in 

Chapter 5. 

Step 5: Report the results of the inspection and evaluation process to the building official and 

building owner. 

4.4.1.1 Detailed Connection Inspections 

In order to perform a detailed inspection of beam-column joints, it is necessary to remove any 

fireproofing or other obscuring finishes to allow direct visual observation of the connection area. 

Detailed inspections may be conducted in stages. An initial stage inspection may be performed 

by removing only the limited amount of fireproofing indicated in Figure 4-1 and following the 

inspection checklist of Section 4.4.1.1.1. If such initial inspection indicates the presence or 

potential presence of damage, than a complete inspection, in accordance with the checklist of 

Section 4.4.1.1.2 should be performed at each connection where such damage is detected. To 

accommodate a complete inspection, removal of fireproofing as indicated in Figure 4-2 is 

necessary. At the discretion of the engineer, a complete inspection in accordance with Section 

4.4.1.1.2 may be performed without first performing the initial inspection of Section 4.4.1.1.1. 

Refer to Chapter 3 for cautions with regard to removal of fireproofing materials. 

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6ˆ

6ˆ

6ˆ

Fireproofing

Exposed surfaces

6ˆ

Figure 4-1 Fireproofing Removal for Initial Connection Inspection

6ˆ

6ˆ

12ˆ

Fireproofing

Figure 4-2 Fireproofing Removal for Complete Connection Inspection

The findings of detailed inspections of moment-resisting connections should be recorded on

appropriate forms, documenting the location of the connection, the person performing the

inspection, the date of the inspection, the extent of the inspection, the means of inspection (visual

or nondestructive testing), the location and type of any observed damage, and, if no damage was

observed, an indication of this. C includes forms suggested for this purpose. 

damage should be classified in accordance with the system of Chapter 2.

Commentary: The largest concentration of reported damage following the 1994

Northridge earthquake occurred at the welded joint between the bottom girder

flange and column, or in the immediate vicinity of this joint. To a much lesser

extent, damage was also observed in some connections at the joint between the

top girder flange and column. 

substantial, then damage is also possible in the panel zone or shear tab areas.

For this reason, and to minimize inspection costs, these Recommended Criteria

Appendix Detected

If damage at either of these locations is
Recommended Postearthquake Evaluation 

FEMA-352 and Repair Criteria for Welded 

Chapter 4: Level 1 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

suggest that it is appropriate to initially inspect only the welded joint of the 

bottom beam flange to the column, and only if damage is found at this location to 

extend the inspection to the remaining connection components. 

4.4.1.1.1 Initial Inspections 

The checklist below may be used as a guide for initial inspections. Prior to performing the 

inspection, remove fireproofing (see Section 3.3.3), as indicated in Figure 4-1. If there are 

indications of damage, then perform a complete inspection in accordance with the procedures of 

Section 4.4.1.1.2. 

� Observe the beam framing into the connection for trueness to line, and potential indications 

of lateral flexural-torsion buckling (damage type G8, Section 2.2.1). 

� Observe condition of fireproofing along the beam within one beam depth of the column for 

cracking or spalling of the fireproofing material along the beam surface, indicating potential 

yielding or buckling of the beam flanges (damage types .G1, G2, Section 2.2.1). 

� Observe the top and bottom surface of the exposed beam bottom flange for fractures (damage 

types G3, G4, Section 2.2.1). 

� Observe the exposed surfaces of the complete joint penetration weld between the beam 

bottom flange and column for fractures (damage types W2, W3, W4 Section 2.2.3). 

� Observe the exposed surfaces of the column flange for fractures (damage types C1, C2, C3, 

Section 2.2.2). 

� Observe the condition at the bottom of weld backing on the bottom flange. If gaps are 

present, insert feeler gauge to detect potential damage (damage types C1, C4, C5, Section 

2.2.2). 

� Observe the bottom surface of the top flange fireproofing at the locations where the beam 

flanges join the column flanges (or continuity plates for minor axis connections) for cracks or 

losses of fireproofing material that could indicate cracking at the complete joint penetration 

weld (damage types G3, Section 5.3.1; C1, C3 and C4 Section 2.2.2; W2, W3, W4, Section 

2.2.3). 

� Observe the condition of the fireproofing at the beam web, in the vicinity of the connection 

from the beam web to the column for loosened, cracked or spalled material indicative of 

potential damage to shear tabs (damage types S1 through S5, Section 2.2.4). 

� Observe the condition of the fireproofing at the column panel zone for cracks, loosened or 

spalled material, indicative of damage to the panel zone or continuity plates (damage types 

P1 through P8, Section 2.2.5). 

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� Observe the flanges of the column at and beneath the joint with the beam flange for loosened, 

spalled or cracked material, indicative of buckled or yielded sections (damage type C6, 

Section 2.2.2). 

4.4.1.1.2 Detailed Inspections 

When an initial inspection conducted in accordance with Section 4.4.1.1.1 indicates the 

presence or likely presence of damage in a connection, the more detailed inspections and 

observations indicated in the checklist below should be performed for that connection. Prior to 

performing the inspection, remove fireproofing (see Section 3.3.3), as indicated in Figure 4-2. 

Note that inspection of the top surface of the top flange of the beam and the adjacent column 

flange will typically be obscured by the diaphragm. If inspections from the exposed bottom 

surface of the top beam flange indicate a potential for damage to be present, then the diaphragm 

should be locally removed to allow a more thorough inspection. 

� Observe the beam framing into the connection for trueness to line, and potential indications 

of lateral flexural-torsion buckling (damage type G8, Section 2.2.1). 

� Observe condition of fireproofing along the beam within one beam depth of the column for 

cracking or spalling of the fireproofing material along the beam surface, indicating potential 

yielding or buckling of the beam flanges (damage types G1, G2, Section 2.2.1). 

� Observe the top and bottom surface of the exposed beam bottom flange and the bottom 

surface of the top flange for fractures (damage types G3, G4, Section 2.2.1). 

� Observe the exposed surfaces of the complete joint penetration welds between the beam top 

and bottom flanges and column for fractures (damage types W2, W3, W4 Section 2.2.3). 

� Observe the exposed surfaces of the column flanges for fractures (damage types C1, C2, C3, 

Section 2.2.2). 

� Observe the condition at the bottom of weld backing on the top and bottom flanges. If gaps 

are present, insert feeler gauge to detect potential damage (damage types C1, C4, C5, Section 

2.2.2). See Chapter 2 for additional information. 

� Observe the condition of the shear tab for deformation of the tab, fractures or tearing of the 

welds and loosening or breaking of the bolts (damage types S1 through S5, Section 2.2.4). 

� Observe the column panel zone for cracks, or distortion (damage types P1 through P8, 

Section 2.2.5). 

� Observe the exposed flanges of the column for distortion (damage type C6, Section 2.2.2). 

4.4.1.2 Damage Characterization 

Characterize the observed damage at each of the inspected connections by assigning a 

connection damage index, dj, obtained either from Table 4-1a or Table 4-1b. Table 4-1a presents 

damage indices for individual classes of damage. Table 4-1b provides indices for the more 

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common combinations of damage and also provides a method for developing indices for other 

combinations. 

Commentary: The connection damage indices provided in Table 4-1 (ranging 

from 0 to 4) represent judgmental estimates of the relative severity of the various 

types of damage. Damage severity is judged in two basic respects, the impact of 

the damage on the connection’s ability to participate in the frame’s global 

stability and lateral resistance, and the impact of the damage on the local gravity 

load carrying capacity of the individual connection. An index of 0 indicates no 

impact on either global or local stability while an index of 4 indicates very severe 

impact. 

When initially developed, in support of the publication of FEMA-267, these 

connection damage indices ranged from 0 to 10 and were conceptualized as 

estimates of the connection’s lost capacity to reliably participate in the building’s 

lateral-force-resisting system in future earthquakes (with 0 indicating no loss of 

capacity and 10 indicating a complete loss of capacity). However, due to the 

limited data available, no direct correlation between these damage indices and 

the actual residual strength and stiffness of a damaged connection was possible. 

In these Recommended Criteria, the damage indices have been simplified, to 

remove the apparent accuracy implied by a scale ranging from 0 to 10. It should 

be noted that although the damage indices do not correlate directly with the loss 

of strength or stiffness experienced by a connection, they do provide a convenient 

qualitative measure of the extent of damage that various connections in a building 

have experienced. 

Analyses conducted to explore the effect of connection fractures on the global 

behavior of frames have revealed that the loss of a single flange connection (top 

or bottom) at each joint, consistently throughout a moment-resisting frame results 

in only a modest increase in the vulnerability of the structure to developing Pdelta instability and collapse. However, if a number of connections develop 

fractures at both flanges of the beam-column connection, significant increase in 

vulnerability occurs. As a result of this, damage that results in the loss of 

effectiveness of a single flange joint to transfer flexural tension stress is assigned 

a relatively modest damage index of 2, if not combined with other types of damage 

at the connection. Damage types that result in an inability of both flanges to 

transfer flexural demands are assigned a high damage index, of 4, as are types of 

damage that could potentially result in impairment of a column or beam’s ability 

to continue to carry gravity loads. Other types of damage are assigned 

proportionately lower damage indices, depending on the apparent effect of this 

damage on structural stability and load carrying capacity. 

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Table 4-1a Connection Damage Indices 

Type Location Description1 Index dj 

G1 Girder Buckled Flange 2 

G2 Girder Yielded Flange 0 

G3 Girder Top or Bottom Flange fracture in Heat Affected Zone (HAZ) 2 

G4 Girder Top or Bottom Flange fracture outside HAZ 2 

G5 Girder Not used -

G6 Girder Yielding or Buckling of Web 2 

G7 Girder Fracture of Web 4 

G8 Girder Lateral-torsional Buckling 2 

C1 Column Minor column flange surface crack 1 

C2 Column Flange tear-out or divot4 2 

C3 Column Full or partial flange crack outside HAZ 3 

C4 Column Full or partial flange crack in HAZ 3 

C5 Column Lamellar flange tearing 2 

C6 Column Buckled Flange 3 

C7 Column Fractured column splice 4 

W2 CJP weld Crack through weld metal exceeding t/4 2 

W3 CJP weld Fracture at girder interface 2 

W4 CJP weld Fracture at column interface 2 

S1 Shear tab Partial crack at weld to column 2 

S2 Shear tab Crack in Supplemental Weld (beam flanges sound) 1 

S3 Shear tab Fracture through tab at bolt holes 4 

S4 Shear tab Yielding or buckling of tab 3 

S5 Shear tab Damaged, or missing bolts3 2 

S6 Shear tab Full length fracture of weld to column 4 

P1 Panel Zone Fracture, buckle, or yield of continuity plate2 1 

P2 Panel Zone Fracture of continuity plate welds2 1 

P3 Panel Zone Yielding or ductile deformation of web2 0 

P4 Panel Zone Fracture of doubler plate welds2 1 

P5 Panel Zone Partial depth fracture in doubler plate2 1 

P6 Panel Zone Partial depth fracture in web2 3 

P7 Panel Zone Full (or near full) depth fracture in web or doubler plate2 4 

P8 Panel Zone Web buckling2 2 

P9 Panel Zone Fully severed column 4 

Notes To Table 4-1a: 

1. See Figures 2-2 through 2-6 for illustrations of these types of damage. 

2. Panel zone damage should be reflected in the damage index for all moment connections that are attached 

to the damaged panel zone within the assembly. 

3. Missing or loose bolts may be a result of construction error rather than damage. The condition of the 

metal around the bolt holes, and the presence of fireproofing or other material in the holes can provide 

clues to this. Where it is determined that construction error is the cause, the condition should be corrected 

and a damage index of “0” assigned. 

4. Damage type C2 is very similar to type W3, the primary differentiation being the depth of the concave 

fracture surface into the column flange. If the fracture surface is relatively shallow within the column 

flange and does not result in the removal of substantial column flange material, type C2 fractures may be 

classified as type W3 and the corresponding damage index utilized. 

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Table 4-1b Connection Damage Indices for Common Damage Combinations 

Girder, Column 

or Weld Damage 

Shear Tab 

Damage 

Damage 

Index 

Girder, Column or 

Weld Damage 

Shear Tab 

Damage 

Damage 

Index 

G3 or G4 S1 4 C5 S1 4 

S2 3 S2 3 

S3 4 S3 4 

S4 3 S4 3

S5 4 S5 4 

Fracture of 

Girder Top 

or Bottom Flange 

S6 4 

Column Flange 

Tearing 

parallel to 

rolling direction 

S6 4 

C2 S1 4 W2, W3, or W4 S1 4 

S2 3 S2 3 

S3 4 S3 4 

S4 3 S4 3

Column Flange 

Tear-out or 

Divot 

S5 4 

CJP Weld 

Fracture 

S5 4 

S6 4 S6 4 

C3 or C4 S1 4 

S2 4 

S3 4 

S4 4 

S5 4 

Column 

Flange 

Crack 

S6 4 

Note: For other combinations of damage, indices are obtained as follows: 

a. Two types of damage with individual di < 1, Combination di =2 

b. Two types of damage with both individual di > 1, Combination di = 4. 

c. Two types of damage with only one individual di > 2, Combination di = largest individual di +1< 4. 

d. Three types of damage with all di < 1, Combination di =3. 

e. Three types of damage with any di > 2, Combination di =4. 

f. More than three types of damage, Combination di =4. 

4.4.1.3 Determine Damage Index at Each Floor for Each Direction of Response 

Divide the connections in the building into two individual groups. Each group of 

connections should consist of those connections, which are part of frames that provide primary 

lateral-force resistance for the structure in one of two orthogonal building directions. For 

example, one group of connections will typically consist of all those connections located in 

frames that provide north-south lateral resistance, while the second group will be all those 

connections located in frames that provide east-west lateral resistance. 

For each group of connections, determine the value of the damage index for the group at each 

floor, from the equation: 

n d 

Di = 

1 � j (4-1) 

n j=1 4 

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where Di is the floor damage index at floor “i” for the group, 

n is the number of connections in the group at floor level “i,” and. 

dj is the damage index, from Tables 4-1a and 4-1b, for the j

th connection in the group 

at that floor. 

4.4.1.4 Determine Maximum Floor Damage Index 

Determine the maximum floor damage index for the building, Dmax, consisting of the largest 

of the Di values calculated in accordance with the Section 4.4.1.3. 

4.4.1.5 Determine Recommended Recovery Strategies for the Building 

Recommended postearthquake recovery strategies are as indicated in Table 4-2, based on the 

maximum damage index, Dmax, determined in accordance with Section 4.4.1.4. 

Table 4-2 Recommended Repair and Modification Strategies 

Values of Dmax 2 Recommended Strategy Note 

0 < Dmax < 0.5 Repair all connections discovered to have dj > 1 

Dmax> 0.5 A potentially unsafe condition should be deemed to exist unless a 

Level 2 evaluation is performed and indicates that acceptable confidence 

is provided with regard to the lateral stability of the structure. Notify the 

building owner of the potentially unsafe condition. Inspect all 

connections in the building. Repair all connections with dj > 1. 

1 

Notes to Table 4-2: 

1. The determination that an unsafe condition may exist should be maintained until either: 

a. Level 2 analyses indicate that a dangerous condition does not exist, or 

b. recommended repairs are completed for all connections having dj > 2. 

2. See Section 4.4.1.4 

Commentary: Recommendations to close a damaged building to occupancy 

should not be made lightly, as such decisions will have substantial economic 

impact, both on the building owner and tenants. A building should be closed to 

occupancy whenever, in the judgment of the structural engineer, damage is such 

that the building no longer has adequate lateral-force-resisting capacity to 

withstand additional strong ground shaking, or if gravity-load-carrying elements 

of the structure appear to be unstable. 

When a building has been damaged, it is recommended that in addition to 

repair, consideration also be given to upgrade. This is particularly the case when 

damage is severe (computed Dmax exceeding 0.5) and the estimated ground 

shaking that caused the damage is substantially less than that which would be 

used to design the building under currently applicable building codes. In such 

conditions, it can reasonably be expected that the building would not be able to 

reliably resist the levels of ground shaking that could credibly occur at the 

building site. In addition to these basic safety considerations, there are also 

economic reasons to consider upgrading a building concurrently with damage 

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repair. A significant portion of structural upgrade costs are a result of the need 

to move occupants out of construction areas as well as the need to selectively 

demolish and replace building finishes and utilities in areas affected by the work. 

Often the magnitude of such costs required to implement repairs are comparable 

to those that would be incurred in performing an upgrade, permitting improved 

future performance to be attained with relatively little increment in construction 

cost. Structural repair, by itself, will not typically result in substantial reduction 

in the vulnerability of the structure to damage from future earthquakes, while 

selected connection upgrade has the potential to greatly reduce future damage 

and losses. 

A companion document to this publication, FEMA-351 – Recommended 

Seismic Evaluation and Upgrade Criteria for Existing Welded Steel MomentFrame Buildings, provides guidelines for assessing the probable performance of 

steel moment-frame buildings and for designing upgrades to improve this 

performance. 

4.4.2 Method 2 – Inspection of a Sample of Connections 

The following eight-step procedure may be used to determine the condition of the structure 

and to develop occupancy, repair and modification strategies when only a sample of the 

building’s critical connections are inspected: 

Step 1: Categorize the moment-resisting connections in the building into two or more groups 

comprising connections expected to have similar probabilities of being damaged. 

Complete steps 2 through 7 below, for each group of connections. 

Step 2: Determine the minimum number of connections in each group that should be 

inspected and select the specific sample of connections to be inspected. 

Step 3: Inspect the selected sample of connections using the procedures of Section 4.4.1 and 

determine connection damage indices, dj, for each inspected connection. 

Step 4: If inspected connections are found to be seriously damaged, perform additional 

inspections of connections adjacent to the damaged connections. 

Step 5: Determine the average damage index d

avg 

for connections in each group, and then the 

average damage index at a typical floor for each group. 

Step 6: Given the average damage index for connections in each group, determine the 

probability P that, had all connections been inspected, the connection damage index 

for any group, at a floor level, would exceed 0.50, and determine the probable 

maximum floor damage index, D . max 

Step 7: Based on the calculated damage indices and statistics, determine appropriate 

occupancy and structural repair strategies. If deemed appropriate, the structural 

engineer may conduct detailed structural analyses of the building in the as-damaged 

state, to obtain improved understanding of its residual condition and to confirm that 

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the recommended strategies are appropriate or to suggest alternative strategies. 

However, for such analyses to be meaningful, full inspections of all connections are 

required. Procedures for such detailed evaluations are contained in Chapter 5. 

Step 8: Report the results of the inspection and evaluation process to the building official and 

building owner. 

Sections 4.4.2.1 through 4.4.2.7 indicate, in detail, how these steps should be performed. 

Commentary: Following an earthquake, structural engineers and technicians 

qualified to perform these evaluations may be in high demand. Prudent owners 

may want to consider having an investigation plan already developed (Steps 1 

and 2) before an earthquake occurs, and to have an agreement with appropriate 

structural engineering and inspection professionals and organizations to give 

priority to inspecting their buildings rapidly following the occurrence of an 

earthquake. 

4.4.2.1 Evaluation Step 1 — Categorize Connections by Groups 

The welded moment-resisting connections participating in the lateral-force-resisting system 

for the building are to be categorized into a series of connection groups. Each group consists of 

connections expected to behave in a similar manner (as an example, a group may consist of all 

those connections that are highly stressed by lateral forces applied in a given direction). As a 

minimum, two groups of connections should be defined - each group consisting of connections 

that primarily resist lateral movement in one of two orthogonal directions. It may be appropriate 

to define additional groups to account for unique conditions, including building configuration, 

construction quality, member size, grade of steel, or other factors that are likely to result in 

connection behavior substantially different from other connections in the building. Each 

connection in the building, including connections at the roof level, should be uniquely assigned 

to one of the groups, and the total number of connections in each group determined. 

In buildings that have significant torsional irregularity, it may be advisable to define at least 

four groups—one group in each orthogonal direction on each side of an assumed center of 

resistance. 

4.4.2.2 Step 2 — Select Samples of Connections for Inspection 

Assign a unique identifier to each connection within each group. Consecutive integer 

identifiers are convenient to some of the methods employed in this Section. 

For each group of connections, select a representative sample for inspection in accordance 

with either of Methods A or B, below. If the evaluation is being performed to satisfy a 

requirement imposed by the building official, a letter indicating the composition of the groups, 

and the specific connections to be inspected should be submitted to the building official prior to 

the initiation of inspection. The owner or structural engineer may at any time in the investigation 

process elect to investigate more connections than required by the selected method. However, 

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the additional connections inspected may not be included in the calculation of damage statistics 

under Step 4 (Section 4.4.2.4) unless they are selected in adherence to the rules laid out for the 

original sample selection, given below. 

Commentary: The purpose of inspection plan submittal prior to the performance 

of inspections is to prevent a structural engineer, or owner, from performing (1) a 

greater number of inspections and (2) reporting data only on those which provide 

a favorable economic result with regard to building disposition. The building 

official need not perform any action with regard to this submittal other than to 

file it for later reference at the time the structural engineer's evaluation report is 

filed. During the inspection process, it may be decided to inspect additional 

connections to those originally selected as part of the sample. While additional 

inspections can be made at any time, the results of these additional inspections 

should not be included in the calculation of the damage statistics, in Step 5, as 

their distribution may upset the random nature of the original sample selection. If 

the additional connections are selected in a manner that preserves the 

distribution character of the original sample, they may be included in the 

calculation of the damage statistics in Step 5. 

4.4.2.2.1 Method A — Random Selection 

In this method, connections should be selected for inspection such that a statistically 

adequate, random sample is obtained. The minimum number of connections to be inspected for 

each group should be determined in accordance with Table 4-3. For groups containing a 

population of 100 connections or more, the sample size need not exceed 18, unless damage with 

dj ‡1.0 in accordance with Table 4-1a is found in the inspection of these 18 connections. In the 

event that such damage is found in this initial sample, the sample size shall be expanded to the 

full amount shown in Table 4-3, while retaining the random character of the selection. 

The following limitations apply to the selection of specific connections: 

1. Up to a maximum of 20% of the total connections in any sample may be pre-selected as those 

expected by rational assessment to be the most prone to damage. Acceptable criteria to select 

these connections could include: 

• Connections shown by a rational analysis to have the highest demand/capacity ratios or at 

locations experiencing the largest drift ratios. 

• Connections that adjoin significant structural irregularities and which therefore might be 

subjected to high localized demands. These include the following irregularities: 

- re-entrant corners 

- set-backs 

- soft or weak stories 

- torsional irregularities (connections at perimeter columns) 

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- diaphragm discontinuities 

• Connections incorporating the largest size framing elements. 

2. The balance of the sample should be selected randomly from the remaining connections in 

the group, except that up to 10% of the connections in the sample may be replaced by other 

connections in the group to which access may more conveniently be made. 

For buildings designed and constructed following the 1994 Northridge earthquake, and 

conforming to the recommendations contained in Chapter 7 of FEMA-267, or conforming to the 

design recommendations for Special Moment Frames contained in the 1997 or later edition of 

AISC Seismic Provisions, the scope of inspection may be reduced to 1/2 the number of 

connections indicated in Table 4-3. If in the course of this reduced scope of inspection, 

significant structural damage is found (damage to any connection with a damage index dj ‡1.0 

from Table 4-1 (a or b)), then full inspections should be performed, as for buildings with other 

types of connections. 

Table 4-3 Minimum Sample Size for Connection Groups 

Number of 

connections in 

Group1 

Minimum number 

of connections to 

be inspected 

Number of 

Connections in 

Group1 

Minimum number of 

connections to be 

inspected 

6 3 200 30 

10 4 300 40 

15 5 400 50 

20 6 500 60 

30 8 750 75 

40 10 1000 100 

50 12 1250 110 

75 16 1500 125 

100 20 2000 150 

Note: 1. For other connection numbers use linear interpolation between values given, rounding up to the 

next highest integer. 

Commentary: The number of connections needed to provide a statistically 

adequate sample depends on the total number of connections in the group, the 

amount of damage present in the building, and the amount of damage it is 

acceptable not to find. Assuming that damage is randomly distributed within a 

connection group, if no damage is found in a randomly selected inspection sample 

of 18 connections, this indicates at least a 95% level of confidence that less than 

15% of the connections in the group have been damaged for a group of any size. 

For smaller groups of connections, smaller samples will provide similar levels of 

confidence. However, if damage is present within the sample of connections 

selected for a group, then a larger sample size will be required to assure with 

confidence that the percentage of connections within the group that have been 

damaged is within a tolerable level. When implemented in the inspection 

procedures contained in these recommended criteria, the inspection sample sizes 

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specified in Table 4-3 will produce greater than a 95% level of confidence of 

finding damage in groups of connections with 20% or more of the connections 

damaged. Military standard, MIL-STD-105D can be used to determine 

appropriate sample sizes to obtain other levels of confidence or to obtain similar 

levels of confidence for reduced levels of damage, if desired. 

If relatively few connections within a group are inspected, the standard 

deviation for the computed damage index will be large. This may result in 

prediction of excessive damage when such damage does not actually exist. The 

structural engineer may elect to investigate more connections than the minimum 

indicated in order to reduce the standard deviation of the sample and more 

accurately estimate the total damage to the structure. These additional 

inspections may be performed at any time in the investigative process. However, 

care should be taken to preserve the random characteristics of the sample, so that 

results are not biased either by selection of connections in unusually heavy (or 

lightly) damaged areas of the structure. 

It is recognized that in many cases the structural engineer may wish to preselect those connections believed to be particularly vulnerable. However, unless 

these pre-selected connections are fairly well geometrically distributed, a number 

that is more than about 10% of the total sample size will begin to erode the 

validity of the assumption of random selection of the sample. If the structural 

engineer has a compelling reason for believing that certain connections are most 

likely to be damaged, and that more than 10% should be pre-selected on this 

basis, either the alternative approach of Method B should be used, or the 

connections that are believed to have particular vulnerability should be classified 

as an independent group, and treated accordingly. 

It is also recognized that there is often a practical incentive to select 

connections that are in specific unoccupied or more accessible areas. It is 

suggested that no more than 10% of the total sample be composed of connections 

pre-selected for this reason. These connections, rather than having a higher 

disposition for damage, might well have a lower than average tendency to be 

damaged. An excessive number of this type of pre-selected connection would 

quickly invalidate the basic assumption of random selection. It is also recognized 

that during the inspection process conditions will be discovered that make it 

impractical to inspect a particular connection, e.g., the architectural finishes are 

more expensive to remove and replace than in other areas, or a particular tenant 

is unwilling to have their space disturbed. However, as discussed above, not 

more than 10% of the total connections inspected should be selected based on 

convenience. 

There are a number of methods available for determining the randomly 

selected portion of the sample. To do this, each connection in the group 

(excluding pre-selected connections) should be assigned a consecutive integer 

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identifier. The sample may then be selected with the use of computer spread sheet 

programs (many of which have a routine for generation of random integers 

between specified limits), published lists of random numbers, or by drawing of 

lots. 

4.4.2.2.2 Method B - Analytical Selection 

In this method, connections should be selected for inspection in accordance with the 

following criteria: 

1. The minimum number of connections within the group to be inspected shall be indicated in 

Table 4-3. As with Method A, if a randomly selected sample of 18 connections from a group 

is inspected, and found to contain no damage, no further inspections of connections from that 

group are required. 

2. Up to 50% of the connections may be selected based on the results of rational analysis 

indicating those connections most likely to be damaged. 

3. The remaining connections in the group to be inspected are selected such that the sample 

contains connections distributed throughout the building, including upper, middle and lower 

stories. The rules of Section 4.4.2.2.1 should be followed in a general way. 

Prior to initiation of the inspections, the rational analysis and list of connections to be 

inspected should be subjected to a qualified independent third party review. The peer review 

should consider the basis for the analysis, consistency of the assumptions employed, and assure 

that overall, the resulting list of connections to be inspected provides an appropriate sampling of 

the building's connections. 

During the inspection process, up to 10% of the connections in the sample may be replaced 

by other connections to which access may more conveniently be made. Substitution for more 

than 10% of the connection sample may be made provided that the independent third party 

reviewer concurs with the adequacy of the resulting revised sample. 

Commentary: In analyses conducted of damaged buildings, there has been a 

generally poor correlation of the locations of damage and the locations of highest 

demand predicted by analysis. This is primarily attributed to the fact that the 

propensity for a fracture to initiate in a connection is closely related to the 

workmanship present in the welded joints, which tends to be a randomly 

distributed quantity. Moreover, typical analysis methods do not capture the 

complex nonlinear stress state that occurs in actual buildings. However, there 

has been some correlation. Analysis is a powerful tool to assist the structural 

engineer in understanding the expected behavior of a structure, damaged or 

undamaged. The specific analysis procedure used should be tailored to the 

individual characteristics of the building. It should include consideration of all 

building elements that are expected to participate in the building's seismic 

response, including, if appropriate, elements not generally considered to be part 

of the lateral-force-resisting system. The ground motion characteristics used for 

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the analysis should not be less than that required by the building code for new 

construction, and to the extent practical, should contain the spectral 

characteristics of the actual ground motion experienced at the site. Qualified 

independent review is recommended to assure that there is careful consideration 

of the basis for the selection of the connections to be inspected and that a 

representative sample is obtained. 

4.4.2.3 Step 3 — Inspect the Selected Samples of Connections 

4.4.2.3.1 Inspection 

All moment-resisting connections within each sample are to be visually inspected as 

indicated in Section 4.4.1.1.2. Where visual inspection indicates the potential for damage that is 

not clearly visible, further investigation using nondestructive testing should be performed. 

Characterize all damage discovered by visual inspection and nondestructive testing for each 

inspected connection as described in Section 4.4.1.1 An individual data sheet (Appendix C) 

should be filled out for each connection inspection, recording its location and conditions 

observed. In addition, plan and elevation sketches for the building’s structural system should be 

developed and conditions of observed damage recorded on these sketches. 

Commentary: The largest concentration of reported damage following the 1994 

Northridge earthquake occurred at the welded joint between the bottom girder 

flange and column, or in the immediate vicinity of this joint. To a much lesser 

extent, damage was also observed in some buildings at the joint between the top 

girder flange and column. If damage at either of these locations is substantial, 

then damage is also commonly found in the panel zone or shear tab areas. 

For a Level 1 evaluation, these Recommended Criteria permit inspection, by 

visual means, of all of the potential damage areas for a representative sample of 

the connections in the building. Most of the damage reported in buildings 

following the 1994 Northridge earthquake consisted of fractures that initiated at 

the roots of complete joint penetration welds joining beam flanges to column 

flanges, and which then propagated through the weld or base metal, leaving a 

trace that was generally detectable by careful visual examination. Careful visual 

examination requires removal of all obscuring finishes and fireproofing, and 

examination from a range of a few inches. Most fractures are visually evident. 

However, some fractures are rather obscure since deformation of the building 

following the onset of fracture can tend to close up the cracks. In some cases, it 

may be appropriate to use magnifying glasses or other means to verify the 

presence of fractures. If doubt exists as to whether a surface indication is really a 

fracture, magnetic particle testing and other forms of nondestructive examination 

can be used to confirm the presence of a fracture. The surface must be carefully 

cleaned prior to testing. 

Some types of fractures extend from the root of the beam flange weld into the 

column flange and may not be detectable by visual examination. Such fractures, 

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typified by types C3 and C5 (see Section 2.2.2) can only be detected by removal of 

the backing, or by nondestructive testing. Often, when such fractures are present, 

a readily visible gap can be detected between the base of the backing and the 

column flange. Where such indications are present, a feeler gauge should be 

inserted into the gap to determine its depth. If the feeler gauge can be inserted to 

a depth that exceeds the backing thickness, a fracture should be assumed to be 

present. Removal of the backing, or nondestructive testing, or both, will be 

required to confirm the extent of the crack. 

The practice of inspecting a small sample of the total connections present in a 

building, in order to infer the probable overall condition of the structure is 

consistent with that followed by most engineers in the Los Angeles area, following 

the 1994 Northridge earthquake. However, the typical practice following that 

event included the extensive use of ultrasonic testing (UT) in addition to visual 

inspection. This UT revealed a number of apparent conditions of damage at the 

roots of the full penetration welds between beam and column flanges. These 

conditions, which were widespread, were typically reported by testing agencies 

and engineers as damage. This practice was encouraged by the FEMA-267 

guidelines, which classified weld root indications as type W1 “damage”. 

As a result of limitations in the accuracy of ultrasonic testing techniques it 

was often found upon removal of weld backing material to allow repair of these 

root conditions, that the actual condition of the weld root was significantly 

different from that indicated by UT. Sometimes, no flaws at all were found at the 

roots of welds reported to have W1 conditions while in other cases, the size and 

location of actual flaws were found to be significantly different from that 

indicated by the UT. 

In the time since, substantial evidence has been gathered that suggests that 

many of the W1 conditions reported following the 1994 Northridge earthquake 

were not damage, but rather latent construction defects, including slag inclusions 

and lack of fusion that had never been detected during the original construction 

quality control and quality assurance processes. For these reasons, these Recommended Criteria have de-emphasized, relative to the recommendations of FEMA267, the importance of employing NDT in the postearthquake inspection process. 

4.4.2.3.2 Damage Characterization 

The observed damage at each of the inspected connections is characterized by assigning a 

connection damage index, dj obtained either from Table 4-1a or Table 4-1b, of Section 4.4.1.2. 

Table 4-1a presents damage indices for individual classes of damage. Table 4-1b provides 

combined indices for the more common combinations of damage and a rule for combining 

indices where a connection has more than one type of damage. Refer to Chapter 2 for 

descriptions of the various damage types and to Section 4.4.1.2 for commentary relative to these 

damage indices. 

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Frame Elevation Floor Plan 

Damaged moment−resisting connection with 1.0  j < dj < 2.0

Adjacent moment−resisting connection − to be inspected 

Transverse connection to be inspected 

Frame Elevation 

Floor Plan 

Damaged moment−resisting connection with dj ‡ 3 

Adjacent moment−resisting connection to be inspected 

Transverse connection to be inspected 

Recommended Postearthquake Evaluation 

FEMA-352 and Repair Criteria for Welded 

Chapter 4: Level 1 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

4.4.2.4 Step 4 — Inspect Connections Adjacent to Damaged Connections 

Regardless of the method used to select the connection sample, perform additional inspections of moment-resisting connections near connections with significant damage as follows: 

• When a connection is determined to have a damage index within the range 1 < dj < 2, inspect 

all of the moment-resisting connections in that line of framing on both sides of the affected 

column and of the column(s) adjacent to the affected column at that floor level and on the 

affected column at the floor level immediately above and below the damaged connection (See 

Figure 4-3). Also inspect any connections for beams framing into the column in the 

transverse direction at that floor level, at the damaged connection. 

• When a connection is determined to have a damage index dj > 3, inspect all of the momentresisting connections in that line of framing on both sides of the affected column and of the 

column(s) adjacent to the affected column at that floor level and on the affected column at the 

two floor levels immediately above and below the damaged connection (See Figure 4-4). 

Also inspect any connections for beams framing into the column in the transverse direction at 

that floor level at the damaged connection. 

Frame Elevation Floor Plan 

Damaged moment−res j isting connection with 1.0 < dj < 2.0

Adjacent moment−resisting connection − to be inspected 

Transverse connection to be inspected 

Figure 4-3 Inspection of Connections Adjacent to Damaged Connection (1 < dj < 2) 

Floor Plan 

Frame Elevation 

Damaged moment−resisting connection with dj ‡ 3 

Adjacent moment−resisting connection to be inspected 

Transverse connection to be inspected 

Figure 4-4 Inspection of Connections Adjacent to Damaged Connection (dj > 3) 

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Assign damage indices dj per Tables 4-1a and 4-1b to each additional connection inspected. 

If significant damage is found in these additional connections (dj > 1), then inspect the 

connections near these additional connections, as indicated above. Continue this process, until 

one of the following conditions occurs: 

• The additional connection inspections do not themselves trigger more inspections, or 

• All connections in the group have been inspected. In this case, proceed with the evaluation 

of damage indices for this group in accordance with the guidelines of Section 4.4.1.3. 

The results of these added connection inspections, performed in this step, are not included in 

the calculation of average damage index davg in Section 4.4.2.6, but are included in the 

calculation of the maximum likely floor damage index Dmax and the probability of excessive 

damage P in Section 4.4.2.7. 

4.4.2.5 Step 5 — Determine Damage Statistics for Each Group 

For each group of connections, determine the estimated average value of the damage index 

for the group davg and its standard deviation s from the equations: 

n 1 davg = �d j (4-2) 

n j =1 

s = ( )

2 

1 1 

1 �=

- -

n 

j

d j d avg n 

(4-3) 

where n is the number of connections in the original sample selected for inspection under 

Step 2 (Section 4.4.2.2), and 

dj is the damage index, from Tables 4-1a and 4-1b, for the j

th inspected connection in 

the original sample 

The additional connections selected using the procedure of Section 4.4.2.4 (Step 4) are not 

included in the above calculation. 

4.4.2.6 Step 6 — Determine the Probability that the Connections in a Group at a Floor 

Level Sustained Excessive Damage 

In this procedure, the probable maximum floor damage index at a floor Dmax is estimated 

from the damage indices determined for all of the connections actually inspected, including those 

additional connections inspected in accordance with the requirements of Section 4.4.2.4. In 

addition, the probability P that had all connections in the building been inspected, Dmax would 

exceed a value of 0.50, is determined. 

First determine the average floor damage index D and its standard deviation S from the 

equations: 

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d 

D = avg (4-4) 4 

s S = 

k 

(4-5)

4 

where davg is the average connection damage index, computed from Equation 4-2, 

s is the standard deviation of the connection damage index, computed from 

Equation 4-3, 

k is the total number of connections (both inspected and not inspected) in the group 

at a typical floor. 

Second, determine the probability P that the set of connections within the group at any floor 

has a floor damage index that is greater than or equal to 0.50. This may be done by using the 

parameters D and S to calculate a factor b, which represents the number of multiples of the 

standard deviation of a normal distribution above the mean that would be required to exceed 1/2. 

The factor b is calculated from the equation: 

b = ( 1 

2 - D) S (4-6) 

Using the value of b calculated from equation 4-6, determine Pf, from Table 4-4. Pf is the 

probability that if all connections had been inspected, the cumulative damage index at any floor 

would have been found to exceed 0.50. If the probability Pf is high, this strongly suggests the 

possibility that there has been a significant reduction in seismic resisting capacity. 

Next, determine the probability P that if all connections within the group had been inspected, 

the connections within the group on at least one floor (out of q total floors in the group) would 

have been found to have a cumulative damage index of 0.50 or more from the equation: 

q P = 1- (1- Pf ) (4-7) 

Finally, for each floor i in the group for which an inspection has been performed, determine 

the floor damage index Di from the equation: 

1 D = ( ki - mi ) davg �� mi d j (4-8) i + � �� 4ki Łł k j =1 4 i 

where: ki is the total number of connections in the group at floor i 

mi is the number of inspected connections in the group at floor i including the 

additional connections inspected under Step 4 

Take Dmax as the largest of the Di values calculated for each floor of the group. 

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Table 4-4 Pf as a Function of Parameter b 

b Pf b Pf 

-1.2816 0.90 1.2265 0.11 

-0.8416 0.80 1.2816 0.10 

-0.5244 0.70 1.3408 0.09 

-0.2533 0.60 1.4051 0.08 

0.0000 0.50 1.4395 0.075 

0.2533 0.40 1.4758 0.07 

0.5244 0.30 1.5548 0.06 

0.8416 0.20 1.6449 0.05 

0.8779 0.19 1.7507 0.04 

0.9154 0.18 1.8808 0.03 

0.9542 0.17 1.9600 0.025 

0.9945 0.16 2.0537 0.02 

1.0364 0.15 2.1701 0.015 

1.0803 0.14 2.3263 0.01 

1.1264 0.13 3.0962 0.001 

1.1750 0.12 3.7190 0.0001 

Note: Intermediate values of Pf may be determined by linear interpolation 

Commentary: The criterion for damage evaluation used in these Recommended 

Criteria is to assume that a cumulative damage index of 0.50 marks the threshold 

at which a structure may become dangerous. Such a damage index could 

correspond to cases where 1/2 of the connections at a floor level have been 

severely damaged, or cases where all of the connections at a floor level have 

experienced moderate damage, or some combination of these, and therefore 

represents a reasonable point at which to begin serious consideration of a 

building’s residual ability to withstand additional loads. 

Although the actual form of the distribution of the probability of damage for 

an individual connection is not known, as the number of connections increases, 

the distribution of damage for a structure tends to a normal distribution, 

regardless of the form of the distribution for individual connections, by the 

Central Limit Theorem. Therefore, the probability that a damage index of 0.50 

has been exceeded at a floor, in a group with k connections, may be approximated 

by determining how many multiples b times the standard deviation S, when added 

to the mean damage index D, equals 1/2. Or, in equation form : 

D + bS = 0.50 (4-9) 

Solution of this equation for the multiplier b results in the required 

relationship of equation 4-6. 

In spite of the somewhat arbitrary nature of the 0.50 damage index criterion 

and the judgmental nature of the suggested way of testing whether that criteria 

has been exceeded, it is believed that the results of these procedures will lead to 

reasonable conclusions in most cases. However, it is always the prerogative of 

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the responsible structural engineer to apply other rational techniques, such as 

direct analyses of the remaining structural strength, stiffness, and deformation 

capacity as a verification of the conclusions provided by these procedures. 

Particularly in anomalous or marginal cases, such additional checks based on 

engineering judgment are strongly encouraged. 

4.4.2.7 Step 7--Determine Recommended Recovery Strategies for the Building 

Recommended postearthquake recovery strategies are as indicated in Table 4-5, based on the 

calculated damage indices and statistics determined in the previous steps. 

Table 4-5 Recommended Condition Designation and Repair Strategies 

Values of Dmax and P Condition Designation Recommended Strategy 

(Cumulative) 

Note 

P<10% and Dmax<0.2 Green - 3 Repair all connections discovered to 

have dj > 1 

1,2 

10% < P < 25 % or 

0.2 < Dmax< 0.5 

Green - 3 Inspect all connections in the group. 

Repair all connections with dj > 1 

1,2 

P > 25 % or 

Dmax> 0.5 

Red – 2 A potentially unsafe condition should 

be deemed to exist unless a level 2 

evaluation is performed and indicates 

that acceptable confidence is provided 

with regard to the lateral stability of 

the structure. Notify the building 

owner of the potentially unsafe 

condition. Inspect all connections in 

the building. Repair all connections 

with dj > 1. Consider structural 

upgrade. 

3 

Notes to Table 4-5: 

1. Includes damage discovered either as part of Step 2 or Step 3. 

2. If all of the discovered damage is relatively minor (dj <1), at the discretion of the engineer, this 

need not be repaired. However, if some of the discovered damage is significant (dj > 1), all of the 

damage should be repaired. 

3. The determination that an unsafe condition may exist should continue until either: 

a. full inspection reveals that the gravity system is not compromised, and that the damage index 

at any floor does not exceed 0.50, or 

b. level 2 analyses indicate that a dangerous condition does not exist, or 

c. recommended repairs are completed for all connections having dj > 1. 

Commentary: Recommendations to close a damaged building to occupancy 

should not be made lightly, as such decisions will have substantial economic 

impact, both on the building owner and tenants. A building should be closed to 

occupancy whenever, in the judgment of the structural engineer, damage is such 

that the building no longer has adequate lateral-force-resisting capacity to 

withstand additional strong ground shaking, or if gravity-load-carrying elements 

of the structure appear to be unstable. 

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When a building has been damaged, it is recommended that, in addition to 

repair, consideration also be given to upgrade. Refer to the additional 

commentary in Section 4.4.1.5. 

4.4.3 Additional Considerations 

Regardless of the value calculated for the damage indices, in accordance with the previous 

sections, and the recommended actions of Section 4.4.2.7, the engineer should be alert for any 

damage condition that results in a substantial lessening of the ability of the structure as a whole, 

or of any part of the structure to resist gravity loads. Should such a condition be encountered, the 

engineer should inform those with legal standing to take appropriate steps either to limit entry to 

the affected portion(s) of the structure, or to ensure that adequate shoring is provide to prevent 

the onset of partial or total building collapse. 

4.5 Evaluation Report 

Upon completion of a detailed evaluation, the responsible structural engineer should prepare 

a written evaluation report and submit it to the person requesting the evaluation, as well as any 

other parties required by law to receive such a report. In particular, the building official should 

be notified whenever a hazardous condition is determined to exist. The report should directly, or 

by attached references, document the inspection program that was performed, and provide an 

interpretation of the results of the inspection program and a general recommendation as to 

appropriate repair and occupancy strategies. The report should include but not be limited to the 

following items: 

• Building address 

• A narrative description of the building, indicating plan dimensions, number of stories, total 

square feet, occupancy, the type and location of lateral-force-resisting elements. Include a 

description of the grade of steel specified for beams and columns and, if known, the type of 

welding (for example, shielded metal are welding or flux-cored arc welding) present. Indicate 

if moment connections are provided with continuity plates. The narrative description should 

be supplemented with sketches (plans and evaluations) as necessary to provide a clear 

understanding of pertinent details of the building's construction. The description should 

include an indication of any structural irregularities as defined in the Building Code. 

• A description of nonstructural damage observed in the building, especially as relates to 

evidence of the drift or shaking severity experienced by the structure. 

• If a letter was submitted to the building official before the inspection process was initiated 

that indicated how the connections were to be divided into groups and indicating the specific 

connections to be inspected, a copy of this letter should be included. 

• A description of the inspection and evaluation procedures used, including the signed 

inspection forms for each individual inspected connection. 

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• A description, including engineering sketches, of the observed damage to the structure as a 

whole (e.g., permanent drift) as well as at each connection, keyed to the damage types in 

Table 4-1a, photographs should be included for all connections with damage index di >1. 

• Calculations of davg, Di, and Dmax for each group, and if all connections in a group were not 

inspected, Pf and P. 

• A summary of the recommended corrective actions (repair and modification measures) and 

any recommendations on occupancy restrictions. 

The report should include identification of any potentially hazardous conditions which were 

observed, including corrosion, deterioration, earthquake damage, pre-existing rejectable 

conditions, and evidence of poor workmanship or deviations or alterations from the approved 

drawings. In addition, the report should include an assessment of the potential impacts of 

observed conditions on future structural performance and recommendations for remediation of 

any adverse conditions. The report should include the Field Inspection Reports of damaged 

connections, as an attachment, and should bear the seal of the structural engineer in charge of the 

evaluation. 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations 

and Repair Criteria for Welded FEMA-352 

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5. LEVEL 2 DETAILED POSTEARTHQUAKE EVALUATIONS 

5.1 Introduction 

Detailed evaluation is the second step of the postearthquake evaluation process. It should be 

performed for all buildings that are estimated to have experienced potentially damaging ground 

motions, using the screening procedures of Section 3.2 of these Recommended Criteria. As 

detailed evaluation can be a time-consuming process, it is recommended that a preliminary 

evaluation, in accordance with the procedures of Chapter 3, be conducted prior to detailed 

evaluation, to permit rapid identification of those buildings that may have been so severely 

damaged that they pose an immediate threat to life safety. 

Many steel moment-frame buildings damaged in past earthquakes have displayed few 

outward signs of structural or nonstructural damage. Consequently, except for those structures 

which have been damaged so severely that they are obviously near collapse, brief evaluation 

procedures, such as those of Chapter 3, are unlikely to provide a good indication of the extent of 

damage or its consequences. In order to make such determination, it is necessary to perform 

detailed inspections of the condition of critical structural components and connections. If 

structural damage is found in the course of such inspections, it is then necessary to make a 

determination as to the effect of discovered damage on the structure’s ability to resist additional 

loading. Ultimately, decisions as to the significance of damage, whether occupancy should be 

permitted in a building and whether specific types of damage should be repaired must be made 

on the basis of quantitative evaluation and engineering judgement. 

Chapter 4 provides a series of recommended criteria for a detailed evaluation method in 

which occupancy and repair decisions are made based on the calculation of damage indices based 

on the observed distribution of damage in the structure. The distribution of damage is 

determined on the basis of detailed inspections of fracture-susceptible connections. Although it 

is preferred that all fracture-susceptible connections be inspected, the procedures of Chapter 4 

permit inspections to be limited to a representative sample. This chapter provides procedures for 

a detailed evaluation processes based on structural analysis of the damaged structure’s ability to 

resist additional strong ground shaking. In order to perform such an analysis it is necessary to 

inspect all fracture-susceptible connections in the building in order to understand their condition. 

Commentary: The Level 1 evaluation approach of Chapter 4 is based on the 

methodology developed immediately after the 1994 Northridge earthquake and 

first presented in FEMA-267. The Level 2 evaluation approach described in this 

chapter is a more comprehensive analytical approach that is compatible with the 

analytical methodology that forms the basis for design and performance 

evaluation criteria contained in the suite of FEMA/SAC publications on steel 

moment frames. 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

5.2 Data Collection 

Prior to performing a detailed evaluation, the original construction drawings should be 

reviewed (if available) to identify the primary lateral and gravity load-resisting systems, typical 

detailing, presence of irregularities, and other features pertinent to structural performance. 

Pertinent available engineering and geotechnical reports, including any previous damage survey 

reports and current estimates of ground motion intensity for the damage causing event, should 

also be reviewed. Specifications (including the original Welding Procedure Specifications) shop 

drawings, erection drawings, and construction records should be reviewed when available. 

When structural framing information is not available, a comprehensive field study must be 

undertaken to determine the location and configuration of all vertical frames, and the details of 

their construction including member sizes, material properties, and connection configurations. A 

companion publication, FEMA-351 – Recommended Seismic Evaluation and Upgrade Criteria 

for Existing Welded Steel Moment-Frame Buildings, provides procedures for obtaining as-built 

information and determining material properties for steel moment-frame buildings. 

Commentary: It is important to collect data on all framing, whether or not it was 

originally intended as part of the design to participate in the lateral force 

resistance of the structure. Studies have shown that vertical frames provided only 

for gravity load resistance can provide substantial supplemental stiffness and 

strength in steel moment-frame structures and the analytical procedures of this 

chapter include direct consideration of such framing. Data collection should 

obtain sufficient information on this framing, as well as that intended to provide 

the structure’s lateral-force resistance to permit an accurate analytical model of 

the structure to be developed. 

In addition to reviewing available documentation, a complete inspection of all critical 

framing and connections in the building should be undertaken, to determine their condition. 

Connections to be inspected include all fracture-susceptible moment-resisting framing 

connections and column splices. The following connections are considered to be fracturesusceptible: 

• Moment-resisting beam-column connections in which the beams are connected to columns 

using full penetration welds between the beam flanges and column, and in which yield 

behavior is dominated by the formation of a plastic hinge within the beam at the face of the 

column, or within the column panel zone. 

• Splices in the exterior columns of steel moment frames when the splices consist of (1) partial 

penetration groove welds between the upper and lower sections of the column, or (2) bolted 

connections that are incapable of developing the full strength of the upper column in tension. 

Section 4.4.1.1 provides procedures for conducting connection inspections, and for 

classifying and recording any damage found. 

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Commentary: Most welded, moment-resisting beam-column connections 

constructed prior to 1994 will be of the fracture susceptible type described here. 

Following the 1994 Northridge earthquake, guidelines for improved connection 

designs and details were developed and were rapidly adopted throughout the 

western United States, particularly in zones of high seismicity, including 

California, Washington, Utah and Alaska. However, fracture-susceptible 

connections may exist in some post-1994 buildings, particularly those constructed 

in zones of lower seismicity. 

5.3 Evaluation Approach 

In a Level 2 evaluation, inspections are conducted of all critical structural elements and 

connections. An analytical model is then developed for the building representing its strength and 

stiffness in the damaged state and an analysis is performed to provide information on the residual 

capacity of the building to resist additional earthquake loading. The results of the analysis are 

used together with engineering judgement and evaluation of other important factors including the 

nature of the building’s occupancy, the economic and other impacts of loss of building use and/or 

building failure, in order to form an opinion as to appropriate postearthquake disposition for the 

building. Alternative actions that may be appropriate in different situations include: 

• Accept the damage as being stable and not detrimental to future building performance, in 

which case no repair action will be required. 

• Determine that repairs of some or all of the damage must be undertaken to provide an 

acceptable level of risk for long-term occupancy, but that the building remains an acceptable 

risk for occupancy until such time as the repairs are completed. 

• Determine that the building is an unacceptable risk for occupancy until such time as 

temporary stabilization or permanent repair can be undertaken. 

• Determine that the building is an unacceptable risk for occupancy until such time as repair 

and structural upgrade can be undertaken. 

• Determine that the building is an unacceptable risk for occupancy and impractical to repair 

and upgrade, in which case the building should be demolished. 

A number of alternative analytical approaches may be used in support of the formation of 

recommendations for postearthquake building disposition. Individual engineers and building 

officials may choose to use any or perhaps several of these approaches, in support of the 

postearthquake decision making process: 

• Determine the capacity of the damaged building relative to current code requirements. 

In this approach the ability of the damaged building to meet the strength and drift criteria 

specified by the building code for new construction is evaluated. Decisions relative to repair 

and occupancy are triggered based on the extent of compliance of the damaged building with 

new building requirements. For example, if a damaged building provides 90% or more of the 

strength and stiffness required of new buildings, and the damage is stable, i.e., not subject to 

further degradation, then it may be appropriate to accept the damage and conduct no repairs. 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

If the degraded strength or stiffness of the building fall below 50% of that required for a new 

building it may be appropriate to restrict occupancy. It should be noted that the engineer, and 

building official, may select any appropriate “trigger” rules when using this approach. 

• Determine the capacity of the damaged building relative to pre-earthquake conditions. 

In this approach, the amount that the strength and stiffness of the building has degraded as a 

result of the damage incurred relative to the pre-earthquake condition is used as an index to 

guide decisions. For example, if a building retains 90% of the strength and stiffness that 

existed prior to the earthquake, and the damage is stable, than it may be appropriate to accept 

the damage and conduct no repairs. If the degraded strength or stiffness of the building fall 

below 50% of the pre-earthquake values, then it may be appropriate to restrict occupancy. As 

noted above, the engineer and building official may select any appropriate “trigger” rules 

when using this approach. 

• Determine the probability of earthquake-induced collapse of the damaged building. In 

this approach, a direct evaluation of the building’s ability to resist collapse for a defined level 

of ground shaking (or at defined hazard probability) is determined and used as a basis for 

making decisions. For example, if analyses permit a high level of confidence to be developed 

that a damaged building can provide Collapse Prevention performance for ground shaking 

demands with a 10% chance of exceedance in 50 years, and the damage is stable, it may be 

appropriate to accept the damage, without repair. Similarly, if a high degree of confidence 

can not be developed that the damaged building could survive ground shaking demands with 

50% chance of exceedance in 50 years, it may be decided to restrict building occupancy. 

Again, the specific “trigger” rules may be selected based upon the judgement of the engineer 

and building official. 

The recommended criteria of this chapter adopt the last approach indicated above. 

Specifically, a methodology is provided whereby the engineer can determine a level of 

confidence with regard to the ability of the damaged building to resist a repeat of the same 

ground shaking that caused the initial damage, without collapse. If a high degree of confidence is 

obtained that the building could survive such ground shaking without significant risk to life 

safety then the building can remain occupied. If there is low confidence that the building can 

protect life safety in a repeat of the same event, then occupancy restrictions are recommended. 

The basic tool used to implement any of the evaluation approaches described above is a 

structural analysis of the damaged building. In addition to presenting detailed criteria for the 

probabilistic evaluation process, this chapter also provides guidance on modeling of damaged 

structures that can be useful with any analytical approach selected by the engineer in assessing 

appropriate postearthquake actions. 

Commentary: As noted, a number of different criteria have historically been used 

to determine whether a building has sustained so much damage that it should not 

continue to remain occupied. In all of these, the decision to post a building 

against occupancy is based on a finding that the building is likely to endanger life 

safety if subjected to additional strong ground shaking. Approaches that have 

most commonly been used in the past include: 

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• comparison of the building’s residual lateral-force-resisting capacity with 

that specified by the building code for design of new structures, 

• comparison of the building’s residual lateral-force-resisting capacity with 

that which existed prior to the onset of damage, and 

• application of the engineer’s judgment as to the extent which the building 

poses an imminent or extreme hazard. 

Each of these approaches has drawbacks. If a comparison of the building’s 

residual lateral-force-resisting capacity with that specified by the building code is 

used, it will often be found that a building that has not been damaged or has only 

minimal damage falls below the trigger level that indicates a “dangerous” 

condition, just due to the fact that the building was designed to earlier editions of 

the code that had less stringent design criteria. This results in a paradox, in that 

engineers typically do not post buildings as “unsafe”, even if they have low 

calculated lateral-force-resisting capacity, unless they have been severely 

damaged. 

The second approach, in which the computed degradation of a building’s 

lateral-force-resisting capacity is used as the measure of whether or not a 

building should be occupied is somewhat more attractive in that it provides a 

direct measure of the effect of the damage sustained on the safety of the building 

and thereby differentiates low-strength conditions that are a result of original 

design characteristics, as opposed to those resulting from damage. However, this 

approach is also somewhat flawed in that some buildings have significant overstrength and reserve capacity and can sustain substantial reduction in initial 

capacity without becoming hazardous. 

Approaches limited to application of the engineers judgment are attractive to 

many engineers, but inherently arbitrary. Further, different engineers will form 

different judgments as to the hazard that damage has caused in a building and 

will recommend different posting actions. 

Review of statistics of past earthquakes indicates that within the relatively 

brief period of a year or so following a major earthquake in a region, the most 

likely events that the region will experience are of a similar or reduced magnitude 

to the original shock. Therefore, these procedures recommend evaluation of 

damaged structures for their ability to resist collapse (ability to provide Collapse 

Prevention performance) for such an event. For the purposes of accounting for 

variability in the likely locations and magnitudes of major aftershocks, and also 

to permit development of confidence levels for ability of the building to provide 

Collapse Prevention performance, a one-year return period is assumed for an 

arbitrary aftershock, comparable in intensity at the building site to the initial 

shock. Variability in ground motion is somewhat arbitrarily accounted for by 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

assuming a distribution of likely ground shaking at the building site due to such 

an aftershock that has a mean value equal to that which caused the original 

damage and having a coefficient of variation of 0.5. 

The safety evaluation approach presented in this section is intended only for 

use in assessing whether a building should remain occupied while it is repaired, 

based on the probability of collapse during the period immediately following the 

earthquake. It is not intended as a tool for evaluating the adequacy of building 

performance over the longer term of the building’s remaining life. For guidelines 

on such performance evaluations refer to the companion publication, FEMA-351, 

Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded 

Steel Moment-Frame Buildings. 

5.4 Field Inspection 

Prior to performing an analytical evaluation of building safety, a thorough inspection of the 

building should be conducted to determine its condition. This inspection should include visual 

inspection of all critical connections including moment-resisting beam-column connections and 

column splices, supplemented by nondestructive testing where visual inspection reveals the 

fracture-susceptible potential damage that cannot be quantified by visual means alone. Beamcolumn connections should be inspected, and the damage recorded, as indicated in Section 4.4.1. 

Geologic site hazards such as fault rupture, landslide, rock fall, and liquefaction may 

influence the damage in a building and also its future performance. A detailed discussion of 

these hazards is provided in FEMA 273 and should be considered as part of a postearthquake 

evaluation. The structure should be inspected to detect whether differential settlement has 

occurred as differential movement between columns in a frame has the potential to place severe 

demands on the moment connections. 

Commentary: Foundation inspection is typically difficult to accomplish since 

most foundations are buried. In most cases, inspection of foundation condition 

can be performed by observing floors for indications of settlement. Where 

significant settlements are indicated, local excavation to expose the foundation 

condition for inspection should be considered. 

5.5 Material Properties and Condition Assessment 

In order to perform a meaningful evaluation, it is necessary to understand the structure’s basic 

configuration, its condition, and certain basic material properties. Original construction 

documents, including the drawings and specifications, supplemented by damage survey reports, 

prepared in accordance with Chapter 4 of these Recommended Criteria, will provide sufficient 

data for the evaluation of most damaged steel moment-frame buildings, so long as the building 

was actually constructed in accordance with these documents. If no construction documents are 

available, then extensive field surveys may be required to define the structure’s configuration, 

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including the locations of frames, the sizes of framing elements and connection details, as well as 

the materials of construction. 

5.5.1 Material Properties 

The primary material properties required to perform analytical evaluations of a steel momentframe building include the following: 

• yield strength, ultimate tensile strength and modulus of elasticity of steel for the columns in 

the moment frames, 

• yield strength, ultimate tensile strength and modulus of elasticity of steel for the beams in the 

moment frames, 

• ultimate tensile strength and notch toughness of the weld metal in the moment-resisting 

connections, and 

• yield and ultimate tensile strength of bolts in the moment-resisting connections. 

Although structural steel is an engineered material, there can be significant variability in the 

properties of the steel in a building, even if all of the members and connection elements conform 

to the same specifications and grades of material. Exhaustive programs of material testing to 

quantify the physical and chemical properties of individual beams, columns, bolts, and welds are 

not justified and should typically not be performed. It is only necessary to characterize the 

properties of material in a structure on the basis of the likely statistical distributions of the 

properties noted above, with characteristic mean values and coefficients of variation. Knowledge 

of the material specification and grade that a structural element conforms to, and its approximate 

age will be sufficient to define these properties for nearly all evaluations. 

In general, analytical evaluations of global building behavior are performed using expected or 

mean values of the material properties based on the likely distribution of these properties for the 

different grades of material present in the structure. Expected values are denoted in these 

procedures with the subscript “e”. Thus, the expected yield and ultimate tensile strength of steel 

are denoted, respectively, Fye and Fue. Some calculations of individual connection capacities are 

performed using lower-bound values of strength. Where lower-bound strength values are 

required, the yield and tensile strength are denoted as Fy and Fu, respectively. Lower-bound 

strengths are defined as the mean minus two standard deviations, based on statistical data for the 

particular specification and grade. 

If original construction documents, including drawings and specifications are available, and 

indicate in an unambiguous manner the materials of construction to be employed, it will typically 

not be necessary to perform materials testing in a steel moment-frame building. When material 

properties are not clearly indicated on the drawings and specifications, or the drawings and 

specifications are not available, the material grades indicated in Table 5-1 may be presumed. 

Alternatively, a limited program of material sample removal and testing may be conducted to 

confirm the likely grades of these materials. 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

Table 5-1 Default Material Specifications for Steel Moment-Frame Buildings 

Element Type Age of Construction Default Specification 

Beams and Columns 1950-1960 ASTM A7, A373 

1961-1990 ASTM A36 

1990-1998 ASTM A572, Grade 50 

1999 and later ASTM A992 

Bolts 1950-1964 ASTM A307 

1964-1999 ASTM A325 

Weld Filler Metal 1950-1964 E6012 or E70241 

1964-1994 E70T4 or E70T72 

1994-1999 See note 3 

Note 1 Prior to about 1964, field structural welding was typically performed with the Shielded Metal Arc 

Welding (SMAW) process using either E6012 or E7024 filler metal. Neither of these electrode 

classifications are rated for specific notch toughness, though some material placed using these 

consumables may provide as much as 40 ft-lbs or greater notch toughness at typical service 

temperatures. It should be noted that due to other inherent characteristics of the moment resisting 

connection detailing prevalent prior to the 1994 Northridge earthquake, the presence of tough filler 

metal does not necessarily provide for reliable ductile connection behavior. 

Note 2 During the period 1964-1994, the Flux Cored Arc Welding (FCAW) process rapidly replaced the 

SMAW process for field welding in building structures. Weld filler metals typically employed for this 

application conformed either to the E70T4 or E70T7 designations. Neither of these weld filler metals 

are rated for specific notch toughness. 

Note 3 Following the 1994 Northridge earthquake, a wide range of weld filler metals were incorporated in 

steel moment-frame construction. Most of these filler metals had minimum ultimate tensile strengths of 

70ksi and minimum rated toughness of 20 ft-lbs at –20o

F. However, due to the variability of practice, 

particularly in the period 1994-1996, a limited sampling of weld metal in buildings constructed in this 

period is recommended to confirm these properties. 

If sampling is performed, it should take place in regions of reduced stress, such as flange tips 

at ends of simply supported beams, flange edges in the mid-span region of members of steel 

moment frames, and external plate edges, to minimize the effects of the reduced area. If a bolt is 

removed for testing, a comparable bolt should be reinstalled in its place. Removal of a welded 

connection sample must be followed by repair of the connection. When sampling is performed 

to confirm the grades of material present in a structure, mechanical properties should be determined in the laboratory using industry standard procedures in accordance with ASTM A-370. 

For the purpose of analytical evaluation of steel moment-frame buildings, the expected and 

lower bound strength of structural materials shall be taken from Table 5-2, based on the age, 

material specification, and grade of material. 

Commentary: In general, great accuracy in the determination of the material 

properties of structural steel elements in steel moment-frame buildings is neither 

justified nor necessary in order to perform reasonably reliable evaluations of 

building performance. The two most important parameters are the yield strengths 

of the beams and columns and the toughness of the weld metal. 

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Table 5-2 Lower Bound and Expected Material Properties for Structural Steel 

Shapes of Various Grades2 

Yield Strength (ksi) Tensile Strength (ksi) 

Material Specification Year of 

Construction 

Lower 

Bound 

Expected Lower 

Bound 

Expected 

ASTM A7, A373 Pre - 1960 30 35 60 70 

ASTM A36 

Group 1 

Group 2 

Group 3 

Group 4 

Group 5 

1961-1990 

41 51 60 70 

39 47 58 67 

36 46 58 68 

34 44 60 71 

39 47 68 80 

ASTM A242, A440, A441 

Group 1 

Group 2 

Group 3 

Group 4 

Group 5 

1960-1970 

45 54 70 80 

41 50 67 78 

38 45 63 75 

38 45 63 75 

38 45 63 75 

ASTM A572 

Group 1 

Group 2 

Group 3 

Group 4 

Group 5 

1970 – 1997 

47 58 62 75 

48 58 64 75 

50 57 67 77 

49 57 70 81 

50 55 79 84 

A36 and Dual Grade 50 

Group 1 

Group 2 

Group 3 

Group 4 

1990 – 1997 

48 55 66 73 

48 58 67 75 

52 57 72 76 

50 54 71 76 

Notes: 

1 Lower bound values for material are mean minus two standard deviations from statistical data. 

Expected values for material are mean values from statistical data. 

2. For wide-flange shapes produced prior to 1997, indicated values are representative of material 

extracted from the web of the section. 

3. For material conforming to ASTM A992, the values for ASTM A572, Grade 50 may be used. No 

adjustment in values, per note 2, should be taken. 

4. For structural plate, expected strength may be taken as 125% of the minimum specified value. 

Lower-bound strength should be taken as the minimum specified value. 

Commentary: In general, great accuracy in the determination of the material 

properties of structural steel elements in steel moment-frame buildings is neither 

justified nor necessary in order to perform reasonably reliable evaluations of 

building performance. The two most important parameters are the yield strengths 

of the beams and columns and the toughness of the weld metal. 

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Weld Filler Metal 

Welds in most steel moment-frame buildings constructed in the period 1964-

1994 were made with the Flux Cored Arc Welding (FCAW) process, employing 

either E70T4 or E70T7 weld filler metal. This material generally has low notch 

toughness. Precise determination of the notch toughness of individual welds is 

not required in order to predict the probable poor performance of momentresisting connections made with these materials and the detailing prevalent until 

1994. However, if weld metal with significant notch toughness (40 ft-lbs at 

service temperature) has been used in a building, even connections of the type 

typically constructed prior to the 1994 Northridge earthquake can provide some 

limited ductility. It is rarely possible to determine the type of weld filler metal 

used in a building without extraction and testing of samples. Construction 

drawings and specifications typically do not specify the type of weld filler metal to 

be employed and even when they do, contractors may make substitutions for 

specified materials. Welding Procedure Specifications (WPS) for a project, if 

available, would define the type of weld filler metal employed, but these 

documents are rarely available for an existing building. Given the near universal 

use of the FCAW process with low toughness weld filler metal during the period 

1964-1994, sampling of weld metal for buildings constructed in this period is not 

recommended. For buildings constructed prior to 1964, sampling and testing of 

weld filler metal may indicate the presence of weld metal with superior notch 

toughness, which would provide a higher level of confidence that the building 

would be capable of meeting desired performance objectives. Buildings 

constructed prior to 1964 may conservatively be assumed to be constructed using 

weld filler metal with low notch toughness, or samples may be extracted. 

Most buildings constructed after 1996 employ weld filler metals with adequate 

notch toughness to provide ductile connection behavior. Sampling and testing of 

welds for buildings constructed in this period are not, therefore, deemed 

necessary. During the period 1994-96, many different types of weld filler metal 

were employed in buildings. Sampling and testing of weld filler metal in 

buildings of this period may be advisable. 

When it is deemed advisable to verify the strength and notch toughness of 

weld filler metals, it is recommended that at least one weld metal sample be 

obtained and tested for each construction type (e.g., column-splice joint, beamflange-to-column-flange joint). Samples should consist of both local base and 

weld metal, such that the composite strength of the connection can be assessed. If 

ductility is required at or near the weld, the design professional may 

conservatively assume, in lieu of testing, that no ductility is available. 

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Beams and Columns 

The actual strength of beam and column elements in a steel moment-frame 

structure is only moderately important for the performance evaluation of such 

structures. The primary parameter used in these Recommended Criteria to 

evaluate building performance, is the interstory drift induced in the building by 

earthquake ground shaking. Building drift is relatively insensitive to the actual 

yield strength of the beams and columns. However, building interstory drift can 

be sensitive to the relative yield strengths of beams and columns. In particular, 

large interstory drifts can occur in buildings with weak columns and strong 

beams, as such conditions permit the development of a single story mechanism in 

which most of the building deformation is accommodated within the single story. 

During the 1970s and 1980s, it was common practice in some regions for 

engineers to specify beams of A36 material and columns of A572, Grade 50 

material in order to develop economical designs with a strong-column-weakbeam configuration. If the properties of materials employed in a steel momentframe building are unknown, it may be conservatively assumed that the beams 

and columns are of the same specification and grade of material, in accordance 

with the default values indicated in Tables 5-1 and 5-2. However, if it can be 

determined that different grades of material were actually used for beams and 

columns, it may be possible to determine a higher level of confidence with regard 

to the ability of a building to meet desired performance objectives. In such cases, 

it may be appropriate to perform a materials sampling and testing program to 

confirm the material specifications for beams and columns. 

When it is decided to conduct a materials testing program to confirm the 

specification and grade of material used in beams and columns, it is suggested 

that at least two tensile strength coupons should be removed from each element 

type for every four floors. If it is determined from testing that more than one 

material grade exists, additional testing should be performed until the extent for 

each grade has been established. 

Bolts 

Bolt specifications may be determined by reference to markings on the heads 

of the bolts. Where head markings are obscured, or not present, the default 

specifications indicated in Table 5-1 may be assumed. If a more accurate 

determination of bolt material is desired, a representative sample of bolts should 

be extracted from the building and subjected to laboratory testing to confirm the 

material grade. 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

5.6 Structural Performance Confidence Evaluation 

The basic process of postearthquake evaluation, as contained in these procedures, is to 

develop a mathematical model of the damaged structure, and by performing structural analysis, to 

determine the likelihood that the building will resist ground shaking demands that can be 

anticipated to occur during the immediate postearthquake period, without collapse. The 

structural analysis is used to predict the value of various structural response parameters. These 

include: 

• interstory drift, and 

• axial forces on columns and column splices. 

These structural response parameters are related to the amount of damage experienced by 

individual structural components as well as the structure as a whole. These procedures specify 

acceptance criteria (median estimates of capacity) for each of the design parameters indicated 

above. Acceptability of structural performance is evaluated considering both local (element 

level) and global performance. Acceptance criteria have been developed on a reliability basis, 

incorporating demand and resistance factors related to the uncertainty inherent in the evaluation 

process, and variation inherent in structural response and capacity, such that a confidence level 

can be established with regard to the ability of a structure to provide specific performance at 

selected probabilities of exceedance. 

Once an analysis is performed, predicted demands are adjusted by two factors, an analysis 

uncertainty factor ga that corrects the analytically predicted demands for bias and uncertainty 

inherent in the analysis technique, and a demand variability factor g that accounts for other 

sources of variability in structural response. These predicted demands are compared against 

acceptance criteria, which have also been factored, by resistance factors, f, to account for 

uncertainties and variation inherent in structural capacity prediction. If the factored demands are 

less than the factored acceptance criteria (capacities), then the structure is indicated to be capable 

of meeting the desired performance, with at least a mean level of confidence. If the factored 

demands exceed the factored acceptance criteria, then there is less than a mean level of 

confidence that the desired performance will be attained. Procedures are given to calculate the 

level of confidence, based on the ratio of factored demand to factored capacity. If the predicted 

level of confidence is inadequate, then the occupancy of the structure should be suspended until 

such time as the structure can be temporarily shored, and/or repaired, and a suitable level of 

confidence attained. In some cases it may be possible to improve the level of confidence with 

regard to the ability of a building to resist collapse by performing a more detailed analysis. More 

detailed and accurate analyses allow better understanding of the structure’s probable behavior to 

be attained, resulting in modifications to the demand and capacity factors. 

Table 5-3 summarizes the recommended posting condition for a building, as a function of the 

level of confidence determined with regard to the structure’s ability to resist collapse for the level 

of ground shaking likely to be experienced in the immediate postearthquake period. Refer to 

Table 3-2 for information on the recommended actions related to each posting. 

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Table 5-3 Recommended Occupancy Actions, Based on Detailed Evaluation 

Confidence Level of Attaining 

Collapse Prevention Performance 

Recommended 

Occupancy 

Posting 

50% or greater confidence of noncollapse Green-1, Green-2, or Green-3, as 

appropriate 

25% or greater confidence of noncollapse but less than 50% Red-1 

Less than 25% confidence of noncollapse Red-2 

Note: Refer to Table 3-2 for explanation of postings. 

Four alternative analytical procedures are considered by these recommendations, for the 

prediction of building response parameters. These are the same basic procedures contained in 

FEMA-273 and include the Linear Static Procedure (LSP), the Linear Dynamic Procedure (LDP), 

Nonlinear Static Procedure (NSP) and Nonlinear Dynamic Procedure (NDP). Section 5.8 outlines 

these procedures in some detail. The reader is referred to FEMA-273 for additional information 

and discussion. 

Commentary: These Recommended Criteria adopt a Demand and Resistance 

Factor Design (DRFD) model for performance evaluation. This approach is 

similar to the Load and Resistance Factor Design (LRFD) approach adopted by 

the AISC design specifications except that the LRFD provisions are conducted on 

an element basis, rather than structural system basis, and the demands in these 

procedures can be drifts as well as forces and stresses. The purpose of this 

DRFD approach is to quantify the level of confidence associated with estimation 

of a damaged building’s ability to provide Collapse Prevention performance 

given the probable ground shaking that may be experienced in the period 

immediately following a damaging earthquake, taken as one year. 

First, it is necessary to presume a hazard relationship for the site, during the 

immediate postearthquake period. Most strong earthquakes are followed by a 

large number of aftershocks, that decrease in frequency over time. Aftershocks 

typically occur on the same fault on which the main shock occurred, though, 

occasionally, an earthquake on a nearby fault has been triggered by the 

redistribution in crustal strains produced by the main shock. Aftershocks 

typically have less magnitude than the main shock, though there are some 

instances when an aftershock has actually exceeded the first shock. This forces a 

change in the naming of the two shocks, to foreshock and main shock. Generally, 

aftershock activity decays to insignificant levels within a period of approximately 

a year following the main event. 

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The actual motion experienced at a site during aftershock activity is 

dependent on the size of the individual events, their location relative to the site 

and the faulting mechanism of the individual events. It is possible for aftershocks 

to produce stronger motion at a specific site than is experienced in the main 

earthquake. For the purposes of this guideline, it is assumed that the probable 

maximum intensity value for aftershock-induced ground shaking at the building 

site is the same as that experienced in the original damaging earthquake, that the 

variability in this intensity is normally distributed and that it has a coefficient of 

variation of 50%. While these assumptions may not be accurate for any specific 

earthquake, and will be conservative for most earthquakes, they present a 

reasonable planning scenario for postearthquake building safety assessments. 

With the above assumptions in place, together with an estimate of the intensity 

of motion that actually occurred at the site during the damaging earthquake, it is 

possible to construct a hazard curve indicating the annual probability of 

exceeding ground motion of defined intensity at the site. For the purposes of 

evaluations conducted in accordance with these Recommended Criteria, the 

hazard curve is plotted as a function of the spectral response acceleration, Sa, at 

the fundamental period of the damaged building, and the annual probability of 

exceedance for these accelerations. Figure 5-1 presents such a hazard curve, 

with spectral response acceleration normalized to the value actually thought to 

have been experienced in the first damaging earthquake. The primary parameters 

of importance from this hazard curve are the slope of the curve evaluated at Sa 

and the value of Sa itself. 

100 

10-1 

10-2 

10-1 100 101 

Ratio of Spectral Acceleration at Period T 

in Main Shock to that in Repeat Shock 

Annual Probability of Exceedance

Figure 5-1 Presumed Postearthquake Hazard Curve 

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Using the Sa value estimated to have been experienced during the first 

damaging earthquake, a structural analysis is performed to determine the 

maximum interstory drift demand for the damaged structure under a repeat of 

that event, as well as the maximum axial forces on critical columns. These 

demands are factored by a demand variability factor g to account for the 

variation associated with estimation of the character of the ground motion and its 

effect on structural response, and an analysis uncertainty factor ga to account for 

the uncertainty and bias inherent in the selected analytical approach. 

The factored demand, gagD calculated from the analysis represents a mean 

estimate of the probable maximum demand during the immediate postearthquake 

period, given the assumed distribution of ground shaking during this period, as 

represented by the assumed hazard curve. 

These Recommended Criteria also specify median estimates of capacity for 

individual elements and the global structure. These capacities are dependent on 

frame and connection configuration. In addition to capacities, capacity 

reduction, or resistance, factors f that adjust the estimated capacity of the 

structure to a mean value are also provided. 

Once the factored demands and capacities are determined, a factoreddemand-to capacity parameter l is calculated from the equation: 

a l = gg D 

(5-1) fC 

where D and C are respectively, the demand and capacity. The value of l is then 

used directly to determine an associated confidence level for the desired 

performance, based on tabulated values related to the uncertainty inherent in the 

estimation of the building’s demands and capacities. Values of l less than 1.0 

indicate greater than mean confidence of achieving the desired performance. 

Values greater than 1.0 indicate less than mean confidence. 

5.7 Ground Motion Representation 

The damaged structure should be analyzed for ground shaking demands representative of 

those that caused the initial damage. Ground shaking demands should be represented in the form 

of a 5% damped elastic response spectrum or with ground acceleration time-histories, compatible 

with this spectrum as required by the selected analytical procedure. Ground shaking demands 

may be determined by one of the following approaches. 

5.7.1 Instrumental Recordings 

When an actual recording of the ground shaking that caused the damage, obtained from the 

building site, or a nearby site with similar conditions is available, this may be used directly to 

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perform analyses of the damaged structure. The ground acceleration time-history should be 

converted into a smoothed, 5% damped response spectrum, similar in form to the generalized 

response spectrum described in FEMA-273, and completely enveloping the actual response 

spectrum obtained for the acceleration record over the period range 0.5T to 2.0T, where T is the 

computed fundamental period of the damaged structure. If the selected analytical procedure is 

response history analysis, a suite of accelerograms constructed in accordance with the 

recommendations of FEMA-273 and matched to the spectrum, should be used, one of which 

should be the actual site recording. 

Commentary: The best possible estimate of ground shaking experienced at a site 

consists of actual ground motion recordings obtained from a free-field instrument 

located at the building site. Free field instruments are preferable to instruments 

located within the building or another structure as they will not be influenced by 

structural response effects. 

Even in zones of high seismicity, few buildings have strong motion 

instrumentation, so it is highly unlikely that such records will be available for 

most buildings. Recordings of ground shaking obtained from other nearby sites 

may be used providing that the site of the instrument is at a comparable distance 

and azimuth to the fault rupture as the damaged building, and providing that site 

soil conditions are reasonably similar. Site soil conditions may be considered to 

be reasonably similar if they are of the same site class, as defined in FEMA-302, 

the NEHRP Recommended Provisions. 

The intent of postearthquake analyses is not to evaluate the damaged 

building’s response for the actual ground shaking that caused the original 

damage, but rather to evaluate this response for ground shaking likely to be 

experienced in the immediate postearthquake period. As previously discussed, 

this is assumed to be similar, though not identical to that which caused the 

original damage. For this reason, response spectra obtained from actual ground 

motion recordings are smoothed, to approximate a standard Newmark and Hall 

spectrum, as described in FEMA-273. 

5.7.2 Estimated Ground Motion 

When instrumental recordings of the damaging ground shaking, as described in Section 5.7.1 

are not available, an estimated response spectrum for this ground shaking should be constructed. 

These spectra should be constructed as recommended by FEMA 273 except that rather than using 

mapped values for the parameters SS and S1, these parameters should be calculated using standard 

attenuation relationships and appropriate estimates of the magnitude of the damage causing 

event, its distance from the building site, the site soil characteristics, faulting mechanism and 

other parameters required by the attenuation equation. Alternatively, these parameters may be 

estimated based on available recordings of ground shaking from the damage causing event. 

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Acceleration time histories, if required, should be constructed in accordance with the 

recommendations of FEMA-273. 

5.8 Analytical Procedures 

In order to evaluate the performance of a damaged steel moment-frame structure it is 

necessary to construct a mathematical model of the damaged structure that represents its strength 

and deformation characteristics and to conduct an analysis to predict the values of various design 

parameters when it is subjected to design ground motion. This section provides procedures for 

selecting an appropriate analysis procedure and for modeling. General requirements for the 

mathematical model are presented in Section 5.9. 

Four alternative analytical procedures are available. The basic procedures are described in 

detail in FEMA-273. This section provides supplementary guidelines on the applicability of the 

FEMA-273 procedures and also provides supplemental modeling recommendations. The four 

basic procedures are: 

• Linear static procedure – an equivalent lateral force technique, similar, but not identical to 

that contained in the building code provisions 

• Linear dynamic procedure – an elastic, modal response spectrum analysis 

• Nonlinear static procedure – a simplified nonlinear analysis procedure in which the forces 

and deformations induced by monotonically increasing lateral loading is evaluated using a 

series of incremental elastic analyses of structural models that are sequentially degraded to 

represent the effects of structural nonlinearity. 

• Nonlinear dynamic procedure – a nonlinear dynamic analysis procedure in which the 

response of a structure to a suite of ground motion time histories is determined through 

numerical integration of the equations of motion for the structure. Structural stiffness is 

altered during the analysis to conform to nonlinear hysteretic models of the structural 

components. 

Commentary: The purpose of structural analyses performed as part of the 

postearthquake assessment process is to predict the values of key response 

parameters, that are indicative of the structure’s performance, when it is 

subjected to ground motion. Once the values of these response parameters are 

predicted, the structure is evaluated for adequacy (appropriate level of confidence 

of achieving desired performance) using the basic approach outlined in Section 

5.6. 

Analyses conducted in these procedures take a markedly different approach 

than those used in the standard design process under the building code 

requirements. Rather than evaluating the forces and deformations induced in the 

structure under arbitrarily reduced loading levels, these analysis procedures 

attempt to predict, within probabilistically defined bounds, the actual values of 

the important response parameters under the design ground motion. 

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The ability of these procedures to estimate reliably the probable performance 

of the structure is dependent on the ability of the analysis to predict the values of 

these response parameters within acceptable levels of confidence. The linear 

dynamic procedure is able to provide relatively reliable estimates of the response 

parameters for structures that exhibit elastic, or near elastic behavior. The linear 

static procedure inherently has more uncertainty associated with its estimates of 

the response parameters because it less accurately accounts for the dynamic 

characteristics of the structure. The nonlinear static procedure is more reliable 

than the linear procedures in predicting response parameters for structures that 

exhibit significant nonlinear behavior, particularly if they are irregular. 

However, it does not accurately account for the effects of higher mode response. 

If appropriate modeling is performed, the nonlinear dynamic approach is most 

capable of capturing the probable behavior of the real structure in response to 

ground motion; however, there are considerable uncertainties associated even 

with the values of the response parameters predicted by this technique. 

5.8.1 Procedure Selection 

Table 5-4 indicates the recommended analysis procedures for various conditions of structural 

regularity and dynamic properties. Note that structural regularity in these procedures is as 

determined in FEMA-273, rather than as alternatively defined in the building codes. Both 

regularity and dynamic properties shall be as determined for the building in the damaged state. 

5.8.2 Linear Static Procedure (LSP) 

5.8.2.1 Basis of the Procedure 

Linear static procedure analysis of damaged steel moment-frame structures shall be 

conducted in accordance with the FEMA-273 Guidelines, except as specifically noted herein. In 

this procedure, lateral forces are applied to the masses of the structure, and deflections and 

component forces under this applied loading is determined. Calculated internal forces typically 

will exceed those that the building can develop, because anticipated inelastic response of 

components and elements is not directly recognized by the procedure. The predicted interstory 

drifts and column axial forces are evaluated using the procedures of Section 5.10. 

Commentary: The linear static procedure is a method of estimating the response 

of the structure to earthquake ground shaking by representing the effects of this 

response through the application of a series of static lateral forces applied to an 

elastic mathematical model of the building’s stiffness. The forces are applied to 

the structure in a pattern that represents the typical distribution of inertial forces 

in a regular structure responding in a linear manner to the ground shaking 

excitation, factored to account, in an approximate manner, for the probable 

inelastic behavior of the structure. It is assumed that the structure’s response is 

dominated by the fundamental mode and that the lateral drifts induced in the 

elastic structural model by these forces represent a reasonable estimate of the 

actual deformation of the structure when responding inelastically. 

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Table 5-4 Selection Criteria for Analysis Procedure to Achieve Collapse Prevention 

Structural Characteristics Analytical Procedure 

Fundamental 

Period, T 

Regularity Ratio of Column 

to Beam Strength 

Linear 

Static 

Linear 

Dynamic 

Nonlinear 

Static 

Nonlinear 

Dynamic 

T < 3.5Ts 

1 Regular2 Strong Column3 Permitted Permitted Permitted Permitted 

Weak Column3 Not 

Permitted 

Not 

Permitted 

Permitted Permitted 

Irregular2 Any Conditions Not 

Permitted 

Not 

Permitted 

Permitted Permitted 

T > 3.5Ts Regular Strong Column3 Not 

Permitted 

Permitted Not 

Permitted 

Permitted 

Weak Column3 Not 

Permitted 

Not 

Permitted 

Not 

Permitted 

Permitted 

Irregular2 Any Conditions Not 

Permitted 

Not 

Permitted 

Not 

Permitted 

Permitted 

Notes: 

1. Ts is the period at which the response spectrum transitions from a domain of constant response acceleration 

(the plateau of the response spectrum curve) to one of constant spectral velocity. Refer to FEMA-273 or 

FEMA-302 for more information. 

2. Conditions of regularity are as defined in FEMA-273. These conditions are significantly different than those 

defined in FEMA-302. 

3. 2.A structure qualifies as having a strong column condition if, at every floor level, the quantity 

�Mpr c / �Mpr b is greater than 1.0, where SMprc and SMprb are the sum of the expected plastic moment 

strengths of the columns and beams, respectively, that participate in the moment-resisting framing in a given 

direction of structural response. 

In the LSP, the building is modeled with linearly-elastic stiffness and 

equivalent viscous damping that approximate values expected for loading to near 

the yield point. Earthquake demands for the LSP are represented by the static 

lateral forces whose sum is equal to the pseudo lateral load. The magnitude of 

the pseudo lateral load has been selected with the intention that when it is applied 

to the linearly elastic model of the building it will result in displacement 

amplitudes approximating maximum displacements that are expected during the 

ground shaking under evaluation. If the building responds essentially elastically 

to the design earthquake, the calculated internal forces will be reasonable 

approximations of those expected during this ground shaking. If the building 

responds inelastically to the earthquake ground shaking, as will commonly be the 

case, when ground shaking is severe, the internal forces that would develop in the 

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yielding building will be less than the internal forces calculated on an elastic 

basis. 

In addition to global structural drift, the collapse of steel moment-frame 

structures is closely related to inelastic deformation demands on the various 

elements that comprise the structure, such as plastic rotation demands on beamcolumn assemblies and tensile demands on column splices. Linear analysis 

methods do not permit direct evaluation of such demands. However, through a 

series of analytical evaluations of typical buildings for a number of earthquake 

records, it has been possible to develop statistical correlation between the 

interstory drift demands predicted by a linear analysis and the actual inelastic 

deformation demands determined by more accurate nonlinear methods. These 

correlation relationships are reasonably valid for regular structures, using the 

definitions of regularity contained in FEMA-273. Thus, the performance 

evaluation process using Linear Static Procedures (LSP) consists of performing 

the LSP analysis to determine an estimate of interstory drift demands, adjustment 

of these demands with the demand factors, g and ga, and comparison with 

tabulated interstory drift capacities. 

Although performance of steel moment-frame structures is closely related to 

interstory drift demand, there are some failure mechanisms, notably, failure of 

column splices, that are more closely related to strength demand. However, since 

inelastic structural behavior affects the strength demand on such elements, linear 

analysis is not capable of directly predicting these demands, except when the 

structural response is essentially elastic. Therefore, when LSP analysis is 

performed for structures that respond in an inelastic manner, column axial 

demands should be estimated using a supplementary plastic analysis approach. 

Two basic assumptions apply in this evaluation approach. First, that the 

distribution of deformations predicted by an elastic analysis is similar to that 

which will occur in actual nonlinear response; second, that the ratio of computed 

strength demands from an elastic analysis to yield capacities is a relative 

indication of the inelastic ductility demand on the element. These assumptions 

are never particularly accurate but become quite inaccurate for structures that 

are highly irregular and experience large inelastic demands. 

Most damaged structures will behave in a more non-linear manner than will 

undamaged structures, even when subjected to relatively low levels of ground 

shaking. Beam-column connections with fractures at the bottom flange of the 

beam, for example, will behave much like undamaged, fully restrained joints when 

loaded such that the fractured flange is in compression, and will behave much like 

pinned joints when loading produces tension at the bottom flange. Such behavior 

can not be accurately reflected in elastic analysis. In order to minimize the 

potential for analysis inaccuracies to result in overly optimistic estimates of the 

actual response of a damaged structure, these Recommended Criteria suggest 

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what are believed to be conservative modeling assumptions for damaged framing 

elements. However, the uncertainties inherent in the use of linear methods to 

model highly damaged structures are so large that it is recommended they not be 

used for this purpose. 

5.8.2.2 Modeling and Analysis Considerations 

When damage results in a structure having different stiffness and strength for loading applied 

positively along one of the principal axes than it does for loading applied negatively, a separate 

model shall be developed and analysis performed for each direction of loading. 

5.8.2.2.1 Period Determination 

A fundamental period shall be calculated for each of two orthogonal directions of building 

response, using standard methods of modal analysis. The model used for this purpose should 

account for the damage sustained. Where damage results in a significantly different stiffness in 

the positive direction of response relative to the negative direction, separate analyses shall be 

performed for each such response direction. 

Commentary: Modal analysis of a model of the building that includes 

representation of the structural damage is required to determine the building’s 

period. This is because approximate formulae, used, for example, in FEMA-302 

for this purpose, may be inaccurate for damaged structures. 

5.8.2.3 Determination of Actions and Deformations 

5.8.2.3.1 Pseudo Lateral Load 

A pseudo lateral load, given by Equation 5-2, shall be independently calculated for each of 

two orthogonal directions of building response, and applied to a mathematical model of the 

building structure. Where damage results in a significantly different stiffness or strength in the 

positive direction of loading than in the negative direction, separate analyses shall be performed 

for each such response direction. 

V = C1C2 C3SaW (5-2) 

where: 

C1 = modification factor to relate expected maximum inelastic displacements to 

displacements calculated for linear elastic response. C1 may be calculated using 

the procedure indicated in Section 3.3.3.3 in FEMA 273 with the elastic base shear 

capacity substituted for Vy. Alternatively, C1 may be taken as having a value of 

1.0 where the fundamental period of response of the structure, T, is greater than Ts 

and shall be taken as having a value of 2.0 where the fundamental period of the 

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structure is equal to or less than T0. Linear interpolation shall be used to calculate 

C1 for intermediate values of T. 

T0 = period at which the acceleration response spectrum for the site reaches 

its peak value, as indicated in FEMA-302. It may be taken as 0.2Ts. 

TS = characteristic period of the response spectrum, defined as the period 

associated with the transition from the constant spectral response 

acceleration segment of the spectrum to the constant spectral response 

velocity segment of the spectrum as defined in FEMA-302. 

C2 = modification factor to represent the effect of hysteretic pinching on maximum 

displacement response. For steel moment-frame structures the value of C2 shall 

be taken as 1.0. 

C3 = modification factor to represent increased dynamic displacements due to P-D 

effects and stiffness degradation. C3 may be taken from Table 5-5 or alternatively, 

shall be calculated from the equation: 

C3 = 1 + 

5(qi - 0.1) ‡ 1.0 (5-3) T 

where: 

qi = the coefficient determined in accordance with Section 3.2.5.1 of FEMA-273. 

Sa = response spectrum acceleration, at the fundamental period and damping ratio of 

the building in the direction under consideration. 

W = total dead load and anticipated live load as indicated below: 

• in storage and warehouse occupancies, a minimum of 25% of the floor live 

load, 

• the actual partition weight or minimum weight of 10 psf of floor area, 

whichever is greater, 

• the applicable snow load – see FEMA-302, and 

• the total weight of permanent equipment and furnishings. 

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Table 5-5 Modification Factors C3 for Linear Static Procedure 

C3 

Ductile fully-restrained connections 1.2 

Brittle fully-restrained connections 1.4 

Notes: 

• Ductile connections are those connections capable of sustaining at least 0.03 radians, 

median, plastic rotation capacity without fracturing or sustaining significant loss of 

strength. 

• Brittle connections are those connections not qualifying as ductile. Typical 

unreinforced moment-resisting connections in which beam flanges are CJP welded to the 

column, using low notch toughness weld filler metal shall be considered brittle unless 

laboratory data are available to substantiate their capability of behaving as indicated for 

ductile connections. 

Commentary: The pseudo lateral force, when distributed over the height of the 

linearly-elastic analysis model of the structure, is intended to produce calculated 

lateral displacements approximately equal to those that are expected in the real 

structure during the design event. If it is expected that the actual structure will 

yield during the design event, the force given by Equation 5-2 may be significantly 

larger than the actual strength of the structure to resist this force. The 

acceptance evaluation procedures in Section 5.10 are developed to take this into 

account. 

The values of the C3 coefficient contained in Table 5-5 are conservative for 

most structures, and will generally result in calculation of an unduly low level of 

confidence. Use of Equation 5-3 to calculate C3 is one way to improve calculated 

confidence without extensive additional effort, and is recommended. 

5.8.2.3.2 Vertical Distribution of Seismic Forces 

The lateral load Fx applied at any floor level x shall be determined as given in Section 

3.3.1.3B of FEMA-273. 

5.8.2.3.3 Horizontal Distribution of Seismic Forces 

The seismic forces at each floor level of the building shall be distributed according to the 

distribution of mass at that floor level. 

5.8.2.3.4 Determination of Interstory Drift 

Interstory drifts shall be calculated using lateral loads in accordance with this section. 

Factored interstory drift demands, gagdi, at each story “i”, shall be determined by applying the 

appropriate demand variability factor g and analytical procedure uncertainty factor ga obtained 

from Section 5.10. 

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Œ œ 

�

� � � 

�

� � � 

Recommended Postearthquake Evaluation 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

5.8.2.3.5 Determination of Column Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

calculated column forces by the applicable analysis uncertainty factor ga and demand variability 

factor g obtained in Section 5.10.3. Column forces shall be calculated either as: 

1. the axial demands from the unreduced linear analysis, or 

2. the axial demands computed from the equation: 

Ø � n M � � n M � ø 

P' c = –Œ2�

�� pe �� - 2�

�� pe �� œ (5-4) 

o Ł i=x L łL Ł i=x L łR ß 

where: 

n M pe �

Ł i=x L �

�

łL 

= the summation of the expected plastic moment strength (ZFye) divided by 

the span length, L, of all moment-connected beams framing into the left 

hand side of the column, above the level under consideration, and 

n M pe � 

Ł i= x L �

�

ł R 

= the summation of the expected plastic moment strength (ZFye) divided by 

the span length, L, of all moment-connected beams framing into the right 

hand side of the column, above the level under consideration. 

When a column is part of framing that resists lateral forces under multiple directions of 

loading, the Seismic Demand shall be taken as the most severe condition resulting from 

application of 100% of the Seismic Demand computed for any one direction of response with 

30% of the Seismic Demand computed for an orthogonal direction of response. 

Commentary: When determining axial demands on columns using Equation 5-4, 

the value of the quantity 2 

M pe may be reduced for beams with fractured L 

connections, when the direction of response of the structure is such that loading 

tends to open the fracture in tension. For such loading, the Mpe value at the 

fracture may be reduced to 30% of the nominal value calculated for the beam. 

Thus, if a beam has a fracture at one end, rather than using the value 2 

M pe for L 

the axial load contribution from this beam, the quantity 1.3 

M pe could be used, L 

when loading tends to place this fracture in tension. If a beam has fractures at 

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both ends that open in tension simultaneously, the contribution for this beam 

M pe could be reduced to 0.6 

L

5.8.3 Linear Dynamic Procedure (LDP) 

5.8.3.1 Basis of the Procedure 

Linear dynamic procedure analysis of damaged steel moment-frame structures should 

generally be conducted in accordance with the FEMA-273 Guidelines, except as specifically 

noted herein. Coefficients C1, C2, and C3 should be taken as indicated in Section 5.8.2.3.1 of 

these Recommended Criteria. 

Estimates of interstory drift and column axial demands shall be evaluated using the 

applicable procedures of Section 5.10. Calculated displacements and column axial demands are 

factored by the applicable analytical uncertainty factor ga and demand variability factor g, 

obtained from Section 5.10, and compared with factored capacity values. Calculated internal 

forces typically will exceed those that the building can sustain because of inelastic response of 

components and elements, but are generally not used to evaluate performance. 

Commentary: The linear dynamic procedure is similar in approach to the linear 

static procedure, described in the previous section. However, because it directly 

accounts for the stiffness and mass distribution of the structure in calculating the 

dynamic response characteristics, it is somewhat more accurate. Coefficients C1, 

C2, and C3, which account in an approximate manner for the differences between 

elastic predictions of drift response and inelastic behavior are the same as for the 

linear static method. Under the Linear Dynamic Procedure (LDP), design 

seismic forces, their distribution over the height of the building, and the 

corresponding internal forces and system displacements are determined using a 

linearly-elastic, dynamic analysis. Note that although the LDP is more accurate 

than the LSP for analysis purposes, it can still be quite inaccurate when applied 

to heavily damaged structures and should be used with caution. 

The basis, modeling approaches, and acceptance criteria of the LDP are 

similar to those for the Linear Static Procedure (LSP). The main exception is that 

the response calculations are carried out using modal spectral analysis. Modal 

spectral analysis is carried out using linearly-elastic response spectra that are 

not modified to account for anticipated nonlinear response. As with the LSP, it is 

expected that the LDP will produce displacements that are approximately correct, 

but will produce internal forces that exceed those that would be obtained in a 

yielding building. 

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5.8.3.2 Modeling and Analysis Considerations 

5.8.3.2.1 General 

The Linear Dynamic Procedure (LDP) should conform to the criteria of this section. The 

analysis should be based on appropriate characterization of the ground motion, as described in 

Section 5.7. The LDP should conform to the criteria in Section 3.3.2.2 of FEMA 273. The 

requirement that all significant modes be included in the response analysis may be satisfied by 

including sufficient modes to capture at least 90% of the participating mass of the building in 

each of the building’s principal horizontal directions. Modal damping ratios should reflect the 

damping inherent in the building at deformation levels less than the yield deformation. Except 

for buildings incorporating passive or active energy dissipation devices, or base isolation 

technology, effective damping should be taken as 5% of critical. 

The interstory drift, and other response parameters calculated for each mode, and required for 

evaluation in accordance with Section 5.8.3.3, should be combined by recognized methods to 

estimate total response. Modal combination by either the SRSS (square root of sum of squares) 

rule or the CQC (complete quadratic combination) rule is acceptable. 

Multidirectional excitation effects may be accounted for by combining 100% of the response 

due to loading in direction A with 30% of the response due to loading in direction B; and by 

combining 30% of the response in direction A with 100% of the response in direction B, where A 

and B are orthogonal directions of response for the building. Where damage to the structure 

results in unsymmetrical response in either the A or B directions, then independent analyses 

should be performed with elements modeled to represent the behavior of the structure when 

pushed in the positive and negative senses along either the A or B directions. 

5.8.3.3 Determination of Actions and Deformations 

5.8.3.3.1 Factored Interstory Drift Demand 

Factored interstory drift demand shall be obtained by mulitplying the results of the response 

spectrum analysis by the product of the modification factors, C1, C2, and C3 defined in Section 

5.8.2.3 and by the analytical procedure uncertainty factor ga and demand variability factor g 

obtained from Section 5.10. 

5.8.3.3.2 Determination of Column Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

calculated column forces, as given in Section 5.8.2.3.5, by the applicable analysis uncertainty 

factor ga and demand variability factor g obtained from Section 5.10.3. 

5.8.4 Nonlinear Static Procedure (NSP) 

5.8.4.1 Basis of the Procedure 

Under the Nonlinear Static Procedure (NSP), a model directly incorporating the inelastic 

material and geometric response of the damaged structure is displaced to a target displacement, 

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and resulting internal deformations and forces are determined. The nonlinear load-deformation 

characteristics of individual components and elements of the damaged building are modeled 

directly. The mathematical model of the building is subjected to a pattern of monotonically 

increased lateral forces or displacements until either a target displacement is exceeded or 

mathematical instability occurs. The target displacement is intended to approximate the total 

maximum displacement likely to be experienced by the actual structure, in response to the 

ground shaking anticipated during the immediate postearthquake period. The target 

displacement shall be calculated by the procedure presented in Section 5.8.4.3.1. Because the 

mathematical model accounts directly for effects of material and geometric inelastic response, 

the calculated internal forces will be reasonable approximations of those expected during the 

anticipated ground shaking, presuming that an appropriate pattern of loading has been applied. 

Results of the Nonlinear Static Procedure (NSP) are to be evaluated using the applicable 

acceptance criteria of Section 5.10. Calculated interstory drifts and column and column splice 

forces are factored, and compared directly with factored acceptable values for the applicable 

performance level. 

5.8.4.2 Modeling and Analysis Considerations 

5.8.4.2.1 General 

In the context of these procedures, the Nonlinear Static Procedure (NSP) involves the 

monotonic application of lateral forces, or displacements, to a nonlinear mathematical model of a 

building, until the displacement of the control node in the mathematical model exceeds a target 

displacement. For buildings that are not symmetric about a plane perpendicular to the applied 

lateral loads, such as often occurs in damaged buildings, the lateral loads must be applied in both 

the positive and negative directions, and the maximum forces and deformations used for design. 

The relation between base shear force and lateral displacement of the control node should be 

established for control node displacements ranging to the target displacement dt, given by 

Equation 3-11 of FEMA 273. Postearthquake assessment shall be based on those column forces 

and interstory drifts corresponding to minimum horizontal displacement of the control node 

equal to the target displacement dt. 

Gravity loads should be applied to appropriate components and elements of the mathematical 

model during the NSP. The loads and load combinations should be as follows: 

1. 100% of computed dead loads and permanent live loads should be applied to the model. 

2. 25% of transient floor live loads should be applied to the model, except in warehouse and 

storage occupancies, where the percentage of live load used in the analysis should be based 

on a realistic assessment of the average long term loading. 

The analysis model should be discretized in sufficient detail to represent adequately the loaddeformation response of each component along its length. Particular attention should be paid to 

identifying locations of inelastic action along the length of a component, as well as at its ends. 

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The modeling and analysis considerations set forth in Section 5.9 should apply to the NDP unless 

the alternative considerations presented below are applied. 

Commentary: As with any nonlinear model, the ability of the analyst to detect the 

presence of inelastic behavior requires the use of a nonlinear finite element at the 

assumed location of yielding. The model will fail to detect inelastic behavior when 

appropriately distributed finite elements are not used. However, as an alternative 

to the use of nonlinear elements, it is possible to use linear elements and 

reconfigure the model, for example, by adjusting member restraints, as 

nonlinearity is predicted to occur. For example, when a member is predicted to 

develop a plastic hinge, a linear model can be revised to place a hinge at this 

location. When this approach is used, the internal forces and stresses that caused 

the hinging must be reapplied, as a nonvarying static load. 

5.8.4.2.2 Control Node 

The NSP requires definition of the control node in a building. These procedures consider the 

control node to be the center of mass at the roof of the building. The top of a penthouse should 

not be considered as the roof. The displacement of the control node is compared with the target 

displacement—a displacement that characterizes the effects of earthquake shaking. 

5.8.4.2.3 Lateral Load Patterns 

Lateral loads should be applied to the building in profiles given in Section 3.3.3.2C of FEMA 

273. 

5.8.4.2.4 Period Determination 

The effective fundamental period Te in the direction under consideration should be calculated 

using the force-displacement relationship of the NSP as described in Section 3.3.3.2D of FEMA 

273. 

5.8.4.2.5 Analysis of Three-Dimensional Models 

Static lateral forces should be imposed on the three-dimensional mathematical model 

corresponding to the mass distribution at each floor level. 

Independent analysis along each principal axis of the three-dimensional mathematical model 

is permitted unless multidirectional evaluation is required by Section 3.2.7 in FEMA 273. Refer 

also to Section 5.8.4.3 of these Recommended Criteria.. 

5.8.4.2.6 Analysis of Two-Dimensional Models 

Mathematical models describing the framing along each axis (axis 1 and axis 2, or the 

orthogonal A and B directions) of the building should be developed for two-dimensional 

analysis. The effects of horizontal torsion should be considered as required by Section 3.2.2.2 of 

FEMA-273. 

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5.8.4.3 Determination of Actions and Deformations 

5.8.4.3.1 Target Displacement 

The target displacement dt for buildings with rigid diaphragms at each floor level shall be 

estimated using the procedures of Section 3.3.3.3 of FEMA-273. Actions and deformations 

corresponding to the control node displacement equal to the target displacement shall be used for 

evaluation in accordance with Section 5.10. 

5.8.4.3.2 Diaphragms 

The lateral seismic load on each flexible diaphragm shall be distributed along the span of that 

diaphragm, considering its displaced shape. 

5.8.4.3.3 Factored Interstory Drift Demand 

Factored interstory drift demand shall be obtained by multiplying the maximum interstory 

drift calculated at the target displacement by the analytical uncertainty factor ga and demand 

variability factor g obtained from Section 5.10.2. 

5.8.4.3.4 Factored Column and Column Splice Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

calculated column forces at the target displacement by the analytical uncertainty factor ga and 

demand variability factor g from Section 5.10.3. 

5.8.5 Nonlinear Dynamic Procedure (NDP) 

5.8.5.1 Basis of the Procedure 

Under the Nonlinear Dynamic Procedure (NDP), design seismic forces, their distribution 

over the height of the building, and the corresponding internal forces and system displacements 

are determined using an inelastic response history dynamic analysis. 

The basis, the modeling approaches, and the acceptance criteria of the NDP are similar to 

those for the NSP. The main exception is that the response calculations are carried out using 

Response-History Analysis. With the NDP, the design displacements are not established using a 

target displacement, but instead are determined directly through dynamic analysis using ground 

motion time-histories. Calculated response can be highly sensitive to characteristics of 

individual ground motions; therefore, it is recommended to carry out the analysis with more than 

one ground motion record. Because the numerical model accounts directly for effects of material 

inelastic response, the calculated internal forces will be reasonable approximations of those 

expected during ground shaking. 

Results of the NDP are to be checked using the applicable acceptance criteria of Section 5.10. 

Calculated displacements and internal forces are factored, and compared directly with factored 

acceptable values. 

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5.8.5.2 Modeling and Analysis Assumptions 

5.8.5.2.1 General 

The NDP should conform to the criteria of given in Section 3.3.4.2A of FEMA-273. 

5.8.5.2.2 Ground Motion Characterization 

The earthquake shaking should be characterized by suites of ground motion acceleration 

histories, prepared in accordance with the recommendations of Section 2.6.2 of FEMA-273 and 

corresponding to the ground motion described in Section 5.7 of these Recommended Criteria. A 

minimum of three pairs of ground motion records should be used. Each pair should consist of 

two orthogonal components of ground motion records. 

Consideration of multidirectional excitation effects required by Section 3.2.7 of FEMA-273 

may be satisfied by analysis of a three-dimensional mathematical model using simultaneously 

imposed pairs of earthquake ground motion records along the horizontal axes of the building. 

The effects of torsion should be considered according to Section 3.2.2.2 of FEMA-273. 

5.8.5.3 Determination of Actions and Deformations 

5.8.5.3.1 Response Quantities 

Response quantities should be computed as follows: 

1. If less than seven pairs of ground motion records are used to perform the analyses, each 

response quantity (for example, interstory drift demand, or column axial demand) should be 

taken as the maximum value obtained from any of the analyses. 

2. If seven or more pairs of ground motion records are used to perform the analyses, the median 

value of each of the response quantities computed from the suite of analyses may be used as 

the demand. The median value shall be that value exceeded by 50% of the analyses in the 

suite. 

5.8.5.3.2 Factored Interstory Drift Demand 

Factored interstory drift demand shall be obtained by multiplying the maximum of the 

interstory drifts calculated in accordance with Section 5.8.5.3.1 by the analytical uncertainty 

factor ga and demand variability factor g obtained from Section 5.10.2. 

5.8.5.3.3 Factored Column and Column Splice Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

column forces calculated in accordance with Section 5.8.5.3.1 by the applicable analytical 

uncertainty factor ga and demand variability factor g obtained from Section 5.10.3. 

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5.9 Mathematical Modeling 

5.9.1 Modeling Approach 

In general, a damaged steel frame building should be modeled, analyzed and designed as a 

three-dimensional assembly of elements and components. Although two-dimensional models 

may provide adequate design information for regular, symmetric structures and structures with 

flexible diaphragms, three-dimensional mathematical models should be used for analysis and 

design of buildings with plan irregularity as defined by FEMA-302. 

Two-dimensional modeling, analysis, and evaluation of buildings with stiff or rigid 

diaphragms is acceptable if torsional effects are either sufficiently small to be ignored or 

indirectly captured. 

Vertical lines of moment frames with flexible diaphragms may be individually modeled, 

analyzed, and evaluated as two-dimensional assemblies of components and elements, or a threedimensional model may be used with the diaphragms modeled as flexible elements. 

If linear or static analysis methods are used, it may be necessary to build separate models to 

simulate the behavior of the structure to ground shaking demands in the positive and negative 

response directions, to account for the differing effects of damage in each direction of response. 

Commentary: An inherent assumption of linear seismic analysis is that the 

structure will exhibit the same stiffness and distribution of stresses regardless of 

whether loads are positively or negatively loaded. However, damage tends to 

create non-symmetrical conditions in structures. For example, fracture damage 

at the bottom flange of a beam will result in a substantial reduction in the 

connection’s stiffness under one direction of loading, but will have negligible 

effect for the reverse direction of loading. In order to capture this behavior using 

linear analysis approaches, it is necessary to build two separate models, one in 

which the damage is effective and one in which the damage is not, to simulate the 

separate response in each direction of loading. A similar approach is required 

for nonlinear static analysis, in that the nonlinear behavior will be different, 

depending on the direction of loading. Only nonlinear dynamic analysis is 

capable of accurately simulating the effects of such damage with a single 

analytical model. 

5.9.2 Model Configuration 

The analytical model should include all frames capable of providing non-negligible stiffness 

for the structure, whether or not intended by the original design to participate in the structure’s 

lateral force resistance. The model should accurately account for any damage sustained by the 

structure. Refer to Section 5.9.11 for procedures on modeling damaged connections. 

Commentary: Gravity framing, in which beams are connected to columns with 

either clip angels or single clip plates can provide significant secondary stiffness 

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to a structure and should in general be modeled when performing postearthquake 

assessment analyses. The primary contributor to this added stiffness is the fact 

that the gravity load columns are constrained to bend to the same deflected shape 

as the columns of the moment-resisting frame, through their interconnection by 

the gravity beams which act as struts, and the diaphragms. As a secondary effect, 

the relatively small rigidity provided by the gravity connections provides some 

additional overall frame stiffness. 

5.9.3 Horizontal Torsion 

The effects of actual horizontal torsion must be considered. In the building codes, the total 

torsional moment at a given floor level includes the following two torsional moments: 

• the actual torsion, that is, the moment resulting from the eccentricity between the centers of 

mass at all floors above and including the given floor, and the center of rigidity of the vertical 

seismic elements in the story below the given floor, and 

• the accidental torsion, that is, an accidental torsional moment produced by horizontal offset 

in the centers of mass, at all floors above and including the given floor, equal to a minimum 

of 5% of the horizontal dimension at the given floor level measured perpendicular to the 

direction of the applied load. 

For the purposes of postearthquake evaluation, under these procedures, accidental torsion 

should not be considered. In buildings with diaphragms that are not flexible, the effect of actual 

torsion should be considered if the maximum lateral displacement dmax from this effect, at any 

point on any floor diaphragm, exceeds the average displacement davg by more than 10%. 

Commentary: Accidental torsion is an artificial device used by the building codes 

to account for actual torsion that can occur, but is not apparent in an evaluation 

of the center of rigidity and center of mass in an elastic stiffness evaluation. Such 

torsion can develop during nonlinear response of the structure if yielding 

develops in an unsymmetrical manner in the structure. For example, if the frames 

on the east and west sides of a structure have similar elastic stiffness, the 

structure may not have significant torsion during elastic response. However, if 

the frame on the east side of the structure yields significantly sooner than the 

framing on the west side, then inelastic torsion will develop. Rather than 

requiring that an accidental torsion be applied in the analysis, as do the building 

codes, these Recommended Criteria directly account for the uncertainty related to 

these torsional effects in the calculation of demand and resistance factors. 

5.9.4 Foundation Modeling 

In general, foundations may be modeled as unyielding. Assumptions with regard to the extent 

of fixity against rotation provided at the base of columns should realistically account for the 

relative rigidities of the frame and foundation system, including soil compliance effects, and the 

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detailing of the column base connections. For purposes of determining building period and 

dynamic properties, soil-structure interaction may be modeled as permitted by the building code. 

Commentary: Most steel moment frames can be adequately modeled by assuming 

that the foundation provides rigid support for vertical loads. However, the 

flexibility of foundation systems (and the attachment of columns to those systems) 

can significantly alter the flexural stiffness at the base of the frame. Where 

relevant, these factors should be considered in developing the analytical model. 

5.9.5 Diaphragms 

Floor and roof diaphragms transfer earthquake-induced inertial forces to vertical elements of 

the seismic-force-resisting system. Development of the mathematical model should reflect the 

stiffness of the diaphragms. As a general rule, most floor slabs with concrete fill over metal deck 

may be considered to be rigid diaphragms and floors or roofs with plywood diaphragms should 

be considered flexible. The flexibility of unfilled metal deck, and concrete slab diaphragms with 

large openings should be considered in the analytical model. Mathematical models of buildings 

with diaphragms that are not rigid should be developed considering the effects of diaphragm 

flexibility. 

5.9.6 P-D Effects 

P-D effects, caused by gravity loads acting on the displaced configuration of the structure, 

may be critical in the seismic performance of steel moment-frame structures, particularly for 

damaged structures that may have significant permanent lateral offset as part of the damage. 

The structure should be investigated to ensure that lateral drifts induced by earthquake 

response do not result in a condition of instability under gravity loads. At each story, the quantity 

yi should be calculated for each direction of response, as follows: 

Pd 

yi = ii (5-5 ) V hyii 

where: 

Pi = portion of the total weight of the structure including dead, permanent live, and 

25% of transient live loads acting on all of the columns within story level i, 

Vyi = total plastic lateral shear force in the direction under consideration at story i, 

hi = height of story i, which may be taken as the distance between the centerline of 

floor framing at each of the levels above and below, the distance between the top 

of floor slabs at each of the levels above and below, or similar common points of 

reference, and 

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di = lateral drift in story i, including any permanent drift, from the analysis in the 

direction under consideration, at its center of rigidity, using the same units as for 

measuring hi. 

In any story in which yi is less than or equal to 0.1, the structure need not be investigated 

further for stability concerns. When the quantity yi in a story exceeds 0.1, the analysis of the 

structure should explicitly consider the geometric nonlinearity introduced by P-D effects. When 

yi in a story exceeds 0.3, the structure shall be considered unstable, unless a detailed global 

stability capacity evaluation for the structure, considering P-D effects, is conducted in accordance 

with the procedures of Appendix A. 

For nonlinear procedures, second-order effects should be considered directly in the analysis; 

the geometric stiffness of all elements and components subjected to axial forces should be 

included in the mathematical model. 

Commentary: The values of interstory drift capacity, provided in Section 5.10, 

and the corresponding resistance factors, were computed considering P-D effects 

(FEMA-355F). For a given structure, it is believed that if the value of y is less 

than 0.3 the effects of P-D have been adequately considered by these general 

procedures. For values of y greater than this limit the statistics on frame 

interstory drift capacities contained in Section 5.10 are inappropriate. For such 

frames explicit determination of interstory drift capacities, considering P-D 

effects using the detailed performance evaluation procedures outlined in 

Appendix A is required. 

The plastic story shear quantity, Vyi, should be determined by methods of 

plastic analysis. In a story in which(1) all beam-column connections meet the 

strong column –weak beam criterion,(2) the same number of moment resisting 

bays is present at the top and bottom of the frame, and (3) the strength of momentconnected girders at the top and bottom of the frame is similar, Vyi may be 

approximately calculated from the equation: 

n 

2� M pG j 

Vyi = 

j =1 (5-6) hi 

where: 

MpGj = the plastic moment capacity of each girder “j” participating in the 

moment resisting framing at the floor level on top of the story. For 

girders with damaged connections, the quantity 2MpGi should be 

taken as the sum of the plastic moment capacities at each end of 

the girder, accounting for the effect of damage on connection 

capacity as recommended in Section 5.9.11. 

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n = the number of moment-resisting girders in the framing at the floor 

level on top of the story. 

In any story in which none of the columns meet the strong-column -weakbeam criterion, the plastic story shear quantity Vyi may be calculated from the 

equation: 

n 

2�M pC k 

V =1= 

k (5-7) yi hi 

where: 

MpCk = the plastic moment capacity of each column “k”, participating in 

the moment resisting framing, considering the axial load present 

on the column. 

For other conditions, the quantity Vyi must be calculated by plastic mechanism 

analysis, considering the vertical distribution of lateral forces on the structure. 

5.9.7 Elastic Framing Properties 

The complete axial area of rolled shapes should be used. For built-up sections, the effective 

area should be reduced if adequate load transfer mechanisms are not available. For elements 

fully encased in concrete, the axial stiffness may be calculated assuming full composite action if 

most of the concrete may be expected to remain after additional ground shaking. Composite 

action may not be assumed for strength unless adequate load transfer and ductility of the concrete 

can be assured. 

The shear area of the elements should be based on standard engineering procedures. The 

comments above regarding built-up section, concrete encased elements, and composite floor 

beam and slab, apply. 

The calculation of rotational stiffness of steel beams and columns in bare steel frames should 

follow standard engineering procedures. For components encased in concrete, the stiffness shall 

include composite action, but the width of the composite section should be taken as equal to the 

width of the flanges of the steel member and should not include parts of the adjoining floor slab, 

unless there is an adequate and identifiable shear transfer mechanism between the concrete and 

the steel. 

5.9.8 Nonlinear Framing Properties 

The elastic component properties, should be computed as outlined in Section 5.9.7. 

Appropriate nonlinear moment-curvature and interaction relationships should be used for beams 

and beam-columns to represent the effects of plastification. 

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5.9.9 Verification of Analysis Assumptions 

Each component should be evaluated to determine that assumed locations of inelastic 

deformations are consistent with strength and equilibrium requirements at all locations along the 

component length. Further, each component should be evaluated by rational analysis for 

adequate postearthquake residual gravity load capacity, considering reduction of stiffness caused 

by earthquake damage to the structure. 

Where moments in horizontally-spanning primary components, due to the gravity loads, 

exceed 50% of the expected moment strength at any location, the possibility for inelastic flexural 

action at locations other than components ends should be specifically investigated by comparing 

flexural actions with expected component strengths. Modeling should account for formation of 

flexural plastic hinges away from component ends when this is likely to occur. 

5.9.10 Undamaged Connection Modeling 

Undamaged connections should be modeled in accordance with the following procedures. 

5.9.10.1 Fully Restrained Connections 

Framing connected with typical welded fully restrained moment-resisting connections, such 

as shown in Figure 5-2, should be modeled as indicated herein. 

Figure 5-2 Welded Unreinforced Fully Restrained Connection (pre-1994) 

5.9.10.1.1 Linear Modeling 

Undamaged fully-restrained connections should be modeled using the gross cross section 

properties and assuming rigid attachment between the beams and columns. Modeling may use 

either center-line-to-center-line dimensions for beams and columns, or alternatively, rigid or 

flexible column panel zones may be modeled to offset the ends of the beams and columns from 

the intersection of the center lines of these members. Rigid offsets, used to represent the panel 

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zone, should not exceed 80% of the dimension of the actual panel zone. Panel zone flexibility 

may be directly considered by adding a panel zone element to the model. 

5.9.10.1.2 Nonlinear Modeling 

Prior to developing a mathematical model for nonlinear analysis of beam-column assemblies 

with welded unreinforced fully restrained moment-resisting connections, an analysis should be 

conducted to determine the controlling yield mechanism for the assembly. This may consist of 

flexural yielding of the beam at the face of the column, flexural yielding of the column at the top 

and/or bottom of the panel zone; shear yielding of the panel zone itself, or a combination of these 

mechanisms. Elements capable of simulating the nonlinear behaviors indicated in these analyses 

should be implemented in the model. Regardless of whether or not panel zones are anticipated to 

yield, panel zones should be explicitly modeled. If calculations indicate that panel zones are 

unlikely to yield in shear, panel zones may be modeled as rigid links. If significant yielding is 

indicated to occur, a suitable element that models this behavior should be used. Expected yield 

strengths Fye should be used for all nonlinear elements to indicate the expected onset of nonlinear 

behavior. Flexural strain hardening of beams and columns should be taken as 5% of the elastic 

stiffness, unless specific data indicates a more appropriate value. Panel zones may be assumed to 

strain harden at 20% of their elastic stiffness. 

5.9.10.2 Simple Shear Tab Connections 

This section presents modeling guidelines for the typical single plate shear tab connection 

commonly used to connect beams to columns for gravity loads, when moment-resistance is not 

required by the design. Figure 5-3 presents a detail for this connection. It is characterized by 

rolled wide flange beams connected to either the major or minor axis of wide flange column 

sections. Beam webs are connected to the column with a single plate shear tab, welded to the 

column and bolted to the beam web. A concrete floor slab, or slab on metal deck may be present 

at the top flange of the beam. 

Major Axis of Column Minor Axis of Column 

Figure 5-3 Typical Simple Shear Tab Connection with Slab 

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5.9.10.2.1 Modeling Guidelines - Linear Analysis 

The connection stiffness should be explicitly modeled as a rotational spring that connects the 

beam to the column. The spring stiffness, Kq should be taken as: 

Kq = 28000(dbg - 5.6) (5-8) 

where dbg is the depth of the bolt group, measured center-line-to-center-line of the outermost 

bolts, in inches and Kq is in units of k-inches per radian. In lieu of explicit modeling of the 

connection, beams that frame into columns with simple shear tab connections may be modeled 

with an equivalent rigidity, EIeq taken as: 

1 EI eq = 

6h 1 (5-9) 

+ 2

lb Kq EI b 

where:

E = the modulus of elasticity, kip/square inch

h = the average story height of the columns above and below the beam, inches

Ib = the moment of inertia of the beam, (inches)

4

lb = the beam span center to center of columns, inches

Commentary: The presence of gravity framing, utilizing shear tab connectors, 

can provide substantial stiffening to the steel moment-frame system provided as 

the basic lateral force resisting system. The primary contributor to this added 

stiffness is the fact that the gravity load columns are constrained to bend to the 

same deflected shape as the columns of the moment-resisting frame, through their 

interconnection by the gravity beams, which act as struts, and the diaphragms. 

The flexural restraint on the columns represented by the spring stiffness given by 

Equations 5-8 and 5-9 is a secondary effect but can provide stability for frames at 

large displacements. 

5.9.10.2.2 Modeling Guidelines - Nonlinear Analysis 

The connection should be explicitly modeled as an elastic-perfectly-plastic rotational spring. 

The elastic stiffness of the spring should be taken as given by Equation 5-8. The plastic strength 

of the spring should be determined as the expected plastic moment capacity of the bolt group, 

calculated as the sum of the expected yield strength of the bolts and their distance from the 

neutral axis of the bolt group. 

5.9.11 Damage Modeling 

This section presents procedures for modeling various conditions of damage. In general, 

damage results in anisotropic frame behavior with affected framing exhibiting different hysteretic 

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properties for loading in a positive direction, than it does for loading in the reverse direction. 

Except for nonlinear dynamic analyses, it is generally necessary to utilize multiple models to 

represent these different behaviors, with loading applied in an appropriate direction for each 

model. 

5.9.11.1 Fully Restrained (FR) Connection Damage 

Damaged type FR connections should be modeled in accordance with the guidelines of this 

section. Refer to Chapter 2 for detailed descriptions of the various damage conditions. 

• Connections with any one of type G3, G4, G7, C2, C4, C5, W2, W3, W4, P5, or P6 damage 

at the bottom flange only or the top flange only may be modeled as undamaged for loading 

conditions in which lateral loading will tend to place the fractured surfaces into compression. 

For loading conditions in which the fracture is placed into tension, the connection should be 

modeled as an undamaged simple shear tab connection, per Section 5.9.10.2. 

• Connections with any combination of type G3, G4, G7, C2, C4, C5, W2, W3, W4, P5, or P6 

damage at the top and bottom flanges should be modeled as an undamaged simple shear tab 

connection, per Section 5.9.10.2 for loading in either direction. 

• If any of the above conditions is present in combination with shear tab damage, types S1, S2, 

S3, S4, S5, or S6, then the connection should be modeled as a simple pin connection for both 

directions of loading. 

• Connections with type P7 damage should be modeled as follows. The beam and column 

above the diagonal plane formed by the fracture should be assumed to be rigidly restrained to 

each other. The beam and column below the diagonal plane formed by the fracture should 

similarly be assumed to be rigidly restrained to each other. The two assemblies consisting of 

the rigidly restrained beam-column joint above and below the diagonal fracture should be 

assumed to be unconnected for loading that places the fracture into tension and should be 

assumed to be connected to each other with a “pin” for conditions of loading that place the 

fracture into compression. 

• Connections with type P9 damage and oriented as indicated in Figure 5-4 should be modeled 

with the beams and columns below the fracture surface assumed to be rigidly connected. The 

column above the fracture surface should be assumed to be unconnected for loading that 

places the column into tension and should be assumed to be “pin” connected for loading that 

places the column into compression. If the orientation of type P9 damage is opposite that 

shown in Figure 5-4, then the instructions above for “top” and “bottom” columns should be 

reversed. 

5.9.11.2 Column Damage 

• If a column has type C1 or C3 damage in any flange, the column should be modeled as if 

having a pinned connection (unrestrained for rotation) at that location for loading conditions 

that induce tension across the fracture. The column may be modeled as undamaged for 

loading conditions that produce compression across the fracture surfaces. 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

P9 

Figure 5-4 Type P9 Panel Zone Damage 

• If a column has type C7, column splice fracture damage, it should be assumed to be 

unconnected across the splice for load conditions that place the column in tension and should 

be assumed to have a “pin” connection for load conditions that place the column in 

compression. 

• If a column has type C6, buckling damage of a flange, the buckled length of the column 

should be modeled with a separate element with flexural properties calculated using only 

30% of the section of the buckled element. 

5.9.11.3 Beam Damage 

• Beams that have lateral torsional buckling, type G8, should be modeled with a flexural pin at 

the center of the buckled region. 

• Beams that have type G1, buckling damage of a flange should be modeled with the buckled 

length of the beam represented by a separate element with flexural properties calculated using 

only 30% of the section of the buckled flange. 

5.9.11.4 Other Damage 

Damage other than indicated in Sections 5.9.11.1, 5.9.11.2, or 5.9.11.3 need not be modeled 

unless in the judgment of the engineer, it results in significant alteration of the stiffness or load 

distribution at the connection. In such cases, the engineer should use judgment in developing the 

model such that it accurately reflects the behavior of the damaged elements. 

5.10 Acceptance Criteria and Confidence Evaluation 

A level of confidence with regard to the building’s ability to provide Collapse Prevention 

performance for a repeat of the original damaging ground motion should be determined. Each of 

the parameters in Table 5-6 must be independently evaluated, using the procedures of Section 

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and Repair Criteria for Welded FEMA-352 

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5.10.1 and the parameters and acceptance criteria of Sections 5.10.2, 5.10.3, and 5.10.4. The 

controlling parameter is that which results in the calculation of the lowest confidence for building 

performance. 

Table 5-6 Performance Parameters Requiring Evaluation of Confidence 

Parameter Discussion 

Interstory drift The maximum interstory drift computed for any story of the structure shall be 

evaluated for global and local behaviors. Refer to Section 5.10.2 

Column axial load The adequacy of each column to withstand the calculated maximum 

compressive demand for that column shall be evaluated. Refer to Section 

5.10.3 

Column splice tension The adequacy of column splices to withstand calculated maximum tensile 

demands for the column shall be evaluated. Refer to Section 5.10.4 

5.10.1 Factored-Demand-to-Capacity Ratio 

Confidence level is determined by first evaluating the factored-demand-to-capacity ratio l 

given by the equation: 

a l = gg D (5-10) fC 

where: 

C = capacity of the structure, as indicated in Sections 5.10.2, 5.10.3, and 5.10.4, for 

interstory drift demand, column compressive demand and column splice tensile 

demand, respectively, 

D = calculated demand for the structure, obtained from the structural analysis, 

g = a demand variability factor that accounts for the variability inherent in the prediction 

of demand related to assumptions made in structural modeling and prediction of the 

character of ground shaking as indicated in Sections 5.10.2, 5.10.3, and 5.10.4, for 

interstory drift demand, column compressive demand and column splice tensile 

demand, respectively, 

ga = an analytical uncertainty factor that accounts for bias and uncertainty inherent in the 

specific analytical procedure used to estimate demand as a function of ground shaking 

intensity as indicated in Section 5.10.2, 5.10.3 and 5.10.4, for interstory drift demand, 

column compressive demand and column splice tensile demand, respectively, 

f = a resistance factor that accounts for the uncertainty and variability inherent in the 

prediction of structural capacity as a function of ground shaking intensity, as indicated 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

in Section 5.10.2, 5.10.3 and 5.10.4, for interstory drift demand, column compressive 

demand and column splice tensile demand, respectively, and 

l = a confidence index parameter from which a level of confidence can be obtained. See 

Table 5-7. 

Factored-demand-to-capacity ratio l shall be calculated using Equation 5-10 for each of the 

performance parameters indicated in Table 5-6, which also references the appropriate section of 

this document where the various parameters, ga, g, and f required to perform this evaluation may 

be found. These referenced Sections 5.10.2, 5.10.3, and 5.10.4 also define an uncertainty 

parameter bUT associated with the evaluation of global and local interstory drift capacity, column 

compressive capacity, and column splice tensile capacity, respectively. These uncertainties are 

related to the building’s configuration, the structural framing system (OMF or SMF), the type of 

analytical procedure employed, and the performance level being evaluated. Table 5-7 indicates 

the level of confidence associated with various values of the factored-demand-to-capacity ratio l 

calculated using Equation 5-10, for various values of the uncertainty parameter bUT. Linear 

interpolation between the values given in Table 5-7 may be used for intermediate values of 

factored-demand-to-capacity ratio l and uncertainty bUT. 

Table 5-7 Factored-Demand-to-Capacity Ratios l and Uncertainty bUT, 

for Specific Confidence Levels 

Uncertainty 

Parameter bUT 

Factored-Demand-to-Capacity Ratios l 

0.2 1.43 1.31 1.23 1.16 1.11 1.05 0.99 0.93 0.86 0.79 0.70 

0.3 1.84 1.62 1.47 1.35 1.25 1.16 1.07 0.97 0.85 0.76 0.63 

0.4 2.49 2.10 1.84 1.65 1.49 1.35 1.21 1.06 0.89 0.77 0.59 

0.5 3.54 2.86 2.44 2.12 1.87 1.65 1.43 1.22 0.99 0.82 0.59 

0.6 5.30 4.10 3.38 2.86 2.46 2.12 1.79 1.48 1.14 0.91 0.62 

Confidence Level 10 20 30 40 50 60 70 80 90 95 99 

Commentary: In order to predict structural performance, these procedures rely 

on the application of structural analysis and laboratory test data to predict the 

behavior of real structures. However, there are a number of sources of 

uncertainty inherent in the application of analysis and test data to performance 

prediction. For example, the actual strength of structural materials, the quality of 

individual welded joints, and the amount of viscous damping present is never 

precisely known, but can have impact on both the actual amount of demand 

produced on the structure and its elements, and on the capacity of the elements to 

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resist these demands. If the actual values of all parameters that affect structural 

performance were known, it would be possible to predict accurately both demand 

and capacity. However, this is never the case. In these procedures, confidence is 

used as a measure of the extent to which predicted behavior is likely to represent 

reality. 

The extent of confidence inherent in a performance prediction is related to the 

possible variation in the several factors that affect structural demand and 

capacity, such as stiffness, damping, connection quality, and the analytical 

procedures employed. In this project, evaluations were made of the potential 

distribution of each of these factors and the effect of variation in these factors on 

structural demand and capacity. Each of these sources of uncertainty in 

structural demand and capacity prediction were characterized as part of the 

supporting research for this project, by a coefficient of variation, bU. The 

coefficient bUT is the total coefficient of variation, considering all sources of 

uncertainty. It is used, together with other factors to calculate the demand and 

resistance factors. It is assumed that demand and capacity are lognormally 

distributed relative to these uncertainty parameters. This allows confidence to be 

calculated as a function of the number of standard deviations that the factoreddemand-to-capacity ratio, l, lies above or below a mean value. Table 5-7 

provides a solution for this calculation, using a value of 5.0 for the hazard 

parameter, k, that is representative of the assumed regional seismicity during the 

year following a major earthquake. Further information on this method may be 

found in Appendix A. 

5.10.2 Performance Limited by Interstory Drift Angle 

5.10.2.1 Factored Interstory Drift Angle Demand 

Factored interstory drift demand should be computed as the quantity, ggaD, where the demand 

D is the largest interstory drift in any story, computed from structural analysis, ga is the 

coefficient obtained from Table 5-8, and g is the coefficient obtained from Table 5-9. 

Commentary: Several structural response parameters are used to evaluate 

structural performance. The primary parameter used for this purpose is 

interstory drift. Interstory drift is an excellent parameter for judging the ability of 

a structure to resist P-D instability and collapse. It is also closely related to 

plastic rotation, or drift angle, demand on individual beam-column connection 

assemblies, and therefore a good predictor of the performance of beams, columns 

and connections. For tall slender structures, a significant portion of interstory 

drift is a result of axial elongation (and shortening) of the columns. Although 

modeling of the structure should account for this frame flexibility, that portion of 

interstory drift resulting from axial column deformation in stories below the story 

under consideration should be neglected in determining local connection 

performance. Unfortunately, this portion of the interstory drift must be 

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Recommended Postearthquake Evaluation 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

determined manually as most computer programs do not separately calculate this 

quantity. 

Table 5-8 Interstory Drift Angle Analysis Demand Uncertainty Factors, ga 

Analysis Procedure LSP LDP NSP NDP 

System Characteristic 

Type 1 Connections 

Low Rise (<4 stories) 0.73 0.86 0.91 1.06 

Mid Rise (4-12 stories) 1.05 1.32 1.02 1.19 

High Rise (> 12 stories) 1.37 1.24 1.02 1.17 

Type 2 Connections 

Low Rise (<4 stories) 1.03 1.40 1.35 1.06 

Mid Rise (4-12 stories) 1.25 1.70 1.46 1.11 

High Rise (> 12 stories) 0.96 1.51 1.71 1.17 

Table 5-9 Interstory Drift Angle Demand Variability Factors, g, 

Type 1 and Type 2 Connections 

Building Height g 

Type 1 Connections1 

Low Rise (< 4 stories) 1.6 

Mid Rise (4 stories – 12 stories) 1.4 

High Rise (>12 stories) 2.0 

Type 2 Connections2 

Low Rise (< 4 stories) 1.7 

Mid Rise (4 stories – 12 stories) 2.0 

High Rise (>12 stories) 2.6 

Notes: 

1- Type 1 connections are capable of resisting median total drift angle demands of 0.04 

radians without fracture or strength degradation. 

2- Type 2 connections are capable of resisting median total drift angle demands of 0.01 

radians without fracture or strength degradation. Generally, welded unreinforced 

connections, employing weld metal with low notch toughness, typical of older steel 

moment-frame buildings should be considered to be of this type. 

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and Repair Criteria for Welded FEMA-352 

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5.10.2.2 Factored Interstory Drift Angle Capacity 

Interstory drift capacity may be limited either by the global response of the structure, or by 

the local behavior of beam-column connections. Section 5.10.2.2.1 provides values for global 

interstory drift capacity for regular, well-configured structures. Global interstory drift capacities 

for irregular structures must be determined using the detailed procedures of Appendix A. Section 

5.10.2.2.2 provides procedures for evaluating local interstory drift angle capacity, as limited by 

connection behavior. 

5.10.2.2.1 Global Interstory Drift Angle 

Factored interstory drift capacity, fC, as limited by global response of the building, shall be 

based on the product of the resistance factor f and capacity C, which are obtained from Table 

5-10, for connections with either Type 1 or Type 2 connections . Type 1 connections are capable 

of resisting median total interstory drift angle demands of 0.04 radians without fracturing or 

strength degradation. Type 2 connections are capable of resisting median total interstory drift 

angle demands of 0.01 radian without fracturing or strength degradation. Welded unreinforced 

moment-resisting connections with weld metal with low notch toughness should be considered 

Type 2. Table 5-11 provides values of the uncertainty coefficient bUT to be used with global 

interstory drift evaluation. 

Table 5-10 Global Interstory Drift Angle Capacity and Resistance Factors 

Structure Type Interstory Drift 

Capacity 

Resistance factor 

f 

Type 1 Connections 

Low Rise (< 4 stories) 0.10 0.85 

Mid Rise (4 stories – 12 stories) 0.10 0.75 

High Rise (>12 stories) 0.085 0.60 

Type 2 Connections 

Low Rise (< 4 stories) 0.10 0.75 

Mid Rise (4 stories – 12 stories) 0.079 0.60 

High Rise (>12 stories) 0.057 0.60 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

Table 5-11 Uncertainty Coefficient bUT for Global Interstory Drift Evaluation 

Building 

Height 

Connection Type 

Type 1 Type 2 

Low Rise (3 stories or less) 0.30 0.35 

Mid Rise ( 4 – 12 stories) 0.40 0.45 

High Rise (> 12 stories) 0.50 0.55 

Notes: 1- Value of bUT should be increased by 0.05 for the linear static procedure. 

2- Value of bUT may be reduced by 0.05 for the nonlinear dynamic procedure. 

5.10.2.2.2 Local Interstory Drift Angle 

Factored interstory drift angle capacity, fC, limited by local connection response, shall be 

based on the capacity of the connection, C, and resistance factor, f, obtained from Table 5-12, for 

the connection types present in the building. Table 5-13 provides values of the uncertainty 

coefficient bUT to be used with local interstory drift evaluation 

Table 5-12 Local Interstory Drift Angle Capacity and Resistance Factors 

Connection Type Interstory Drift Capacity Resistance factor f 

Pre-Northridge connection with low notch 

toughness weld metal 

0.053-0.0006db 0.7 

Pre-Northridge connection with notch tough 

weld metal (Note 1) 

0.060-0.0006db 0.85 

Shear tab connections 0.16-0.0036db 0.7 

Post-Northridge connection intended for steel 

moment-frame Service (Note 2) 

0.04 0.85 

Notes: 

1. Weld metal with a notch toughness 40 ft –lbs at anticipated service temperature 

2. Many types of connections approved for steel moment-frame service in the post-Northridge period are 

capable of better performance than this. Refer to FEMA-350, Recommended Seismic Design Criteria 

for New Steel Moment-Frame Buildings for more detailed data. 

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Table 5-13 Uncertainty Coefficient bUT for Local Interstory Drift Evaluation 

Building 

Height 

Connection Type 

Type 1 Type 2 

Low Rise (3 stories or less) 0.30 0.35 

Mid Rise ( 4 – 12 stories) 0.35 0.40 

High Rise (> 12 stories) 0.40 0.40 

Notes: 1- Value of bUT should be increased by 0.05 for linear static analyses. 

2- Value of bUT may be reduced by 0.05 for nonlinear dynamic analyses. 

5.10.3 Performance Limited by Column Compressive Capacity 

5.10.3.1 Column Compressive Demand 

Factored column compressive demand shall be determined for each column as the quantity 

ggaD, where: 

D = the compressive axial load on the column determined as the sum of Dead Load, 25% of 

unreduced Live Load, and Seismic Demand. Seismic Demand shall be determined by 

either of the following four analysis methods: 

Linear: The axial demands may be taken as those predicted by a linear 

static or linear dynamic analysis, conducted in accordance with 

Section 5.8.2 or 5.8.3 of these Recommended Criteria. 

Plastic: The axial demands may be taken based on plastic analysis, as 

indicated by Equation 5-4 of Section 5.8.2.3.5 of these 

Recommended Criteria. 

Nonlinear Static: The axial demands may be taken based on the computed forces 

from a nonlinear static analysis, at the target displacement, in 

accordance with Section 5.8.4 of these Recommended Criteria. 

Nonlinear Dynamic: The axial demands may be taken based on the computed design 

forces from a nonlinear dynamic analysis, in accordance with 

Section 5.8.5 of these Recommended Criteria. 

ga = Analytical demand uncertainty factor, taken from Table 5-14. 

g = Demand variability factor, taken as having a value of 1.1. 

The uncertainty coefficient bUT shall be taken as indicated in Table 5-14 based on the 

procedure used to calculate column compressive demand D. 

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Recommended Postearthquake Evaluation 

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Chapter 5: Level 2 Detailed Postearthquake Evaluations Steel Moment-Frame Buildings 

Table 5-14 Analysis Uncertainty Factor ga and Total Uncertainty Coefficient bUT for 

Evaluation of Column Compressive Demands 

Analytical Procedure Analysis Uncertainty 

Factor ga 

Total Uncertainty 

Coefficient bUT 

Linear static or dynamic analysis 1.15 0.35 

Plastic analysis (Section 4.4.3.3.6) 1.0 0.15 

Nonlinear static analysis 1.05 0.20 

Nonlinear dynamic analysis 2

1.4 

e 

b 2

0.0225 + b

Note: b may be taken as the coefficient of variation of the axial load values determined from the suite of 

nonlinear analyses. 

Commentary: The value of g has been computed assuming a coefficient of 

variation for axial load values resulting from material strength variation and 

uncertainty in dead and live loads of 15%. The values of ga have been calculated 

assuming coefficients of variation of 30%, 0% and 15% related to uncertainty in 

the analysis procedures for linear, plastic and nonlinear static analyses, 

respectively. In reality, for structures that are stressed into the inelastic range, 

elastic analysis will typically overestimate axial column demands, in which case, 

a value of 1.0 could be used. However, for structures that are not loaded into the 

inelastic range, the indicated value is appropriate. Plastic analysis will also 

typically result in an upper bound estimate of column demand, and application of 

additional demand factors is not appropriate. For nonlinear dynamic analysis, 

using a suite of ground motions, direct calculation of the analysis demand factor 

is possible, using the equation shown. All of these demand factors are based on a 

hazard parameter k, having a value of 5.0, representative of the assumed 

seismicity for the immediate postearthquake period. 

5.10.3.2 Column Compressive Capacity 

Factored compressive capacity of each individual column to resist compressive axial loads 

shall be determined as the product of the resistance factor, f, and the nominal axial strength of 

the column, C, which shall be determined in accordance with the AISC Load and Resistance 

Factor Design Specification. Specifically, for the purposes of this evaluation, the effective length 

coefficient k shall be taken as having a value of 1.0 and the resistance factor f shall be assigned a 

value of 0.90. 

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5.10.4 Column Splice Capacity 

The capacity of column tensile splices, other than splices consisting of complete joint 

penetration (CJP) butt welds of all elements of the column (flanges and webs) shall be evaluated 

in accordance with this section. Column splices consisting of CJP welds of all elements of the 

column, and in which the weld filler metal has a minimum notch toughness of 40 ft-lbs at the 

lowest anticipated service temperature, need not be evaluated. 

5.10.4.1 Column Splice Tensile Demand 

Factored column splice tensile demand shall be determined for each column as the quantity 

ggaD, where D is the column splice tensile demand. Column splice tensile demand shall be 

determined as the computed Seismic Demand in the column, less 90% of the computed Dead 

Load demand. Seismic Demand shall be as determined for column compressive demand, in 

accordance with Section 5.10.3.1. The demand variability factor g shall be taken as having a 

value of 1.05 and the analysis uncertainty factor ga shall be taken as indicated in Table 5-14. The 

total uncertainty coefficient bUT shall also be taken as indicated in Table 5-14. 

5.10.4.2 Column Splice Tensile Capacity 

The capacity of individual column splices to resist tensile axial loads shall be determined as 

the product of the resistance factor, f, and the nominal tensile strength of the splice, C, as 

determined in accordance with the AISC Load and Resistance Factor Design Specification. 

Specifically, Chapter J shall be used to calculate the nominal tensile strength of the splice 

connection. For the purposes of this evaluation, f shall be assigned a value of 0.85. 

5.11 Evaluation Report 

Regardless of the level of evaluation performed, the responsible structural engineer should 

prepare a written evaluation report and submit it to the owner upon completion of the evaluation. 

When the building official has required evaluation of a steel moment-frame building, this report 

should also be submitted to the building official. This report should directly, or by attached 

references, document the inspection program that was performed, and provide an interpretation 

of the results of the inspection program and a general recommendation as to appropriate repair 

and occupancy strategies. The report should include but not be limited to the following material: 

• Building address 

• A narrative description of the building, indicating plan dimensions, number of stories, total 

square feet, occupancy, and the type and location of lateral-force-resisting elements. Include a 

description of the grade of steel specified for beams and columns and, if known, the type of 

welding (e.g., Shielded Metal Arc Welding, or Flux-Cored Arc Welding) present. Indicate if 

moment connections are provided with continuity plates. The narrative description should be 

supplemented with sketches (plans and evaluations) as necessary to provide a clear 

understanding of pertinent details of the building’s construction. The description should 

include an indication of any structural irregularities as defined in the Building Code. 

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• A description of nonstructural damage observed in the building. 

• An estimate of the ground shaking intensity experienced by the building, determined in 

accordance with Section 5.7. 

• A description of the inspection and evaluation procedures used, including the signed 

inspection forms for each individual inspected connection. 

• A description, including engineering sketches, of the observed damage to the structure as a 

whole (e.g., permanent drift) as well as at each connection, keyed to the damage types in 

Chapter 5; photographs should be included for all connections with significant visible 

damage. 

• Calculations demonstrating the determination of a confidence level with regard to the 

building’s ability to resist collapse in the immediate postearthquake period. 

• A summary of the recommended actions (repair and modification measures and occupancy 

restrictions). 

The report should include identification of any potentially hazardous conditions that were 

observed, including corrosion, deterioration, earthquake damage, pre-existing rejectable 

conditions, and evidence of poor workmanship or deviations from the approved drawings. In 

addition, the report should include an assessment of the potential impacts of observed conditions 

on future structural performance. The report should include the Field Inspection Reports of 

damaged connections (visual inspection and nondestructive testing records, data sheets, and 

reports), as an attachment, and should bear the seal of the structural engineer in charge of the 

evaluation. 

The nature and scope of the evaluations performed should be clearly stated in 

the structural engineer’s written evaluation report. If the scope of evaluation does 

not permit an informed judgment to be made as to the extent with which the 

building complies with the applicable building codes, or as to a statistical level of 

confidence that the damage has not exceeded an acceptable damage threshold, 

this should be stated . 

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Steel Moment-Frame Buildings Chapter 6: Postearthquake Repair 

6. POSTEARTHQUAKE REPAIR 

6.1 Scope 

This section provides criteria for structural repair of earthquake damage. Repair constitutes 

any measures taken to restore earthquake damaged elements of the building, including individual 

members or their connections, or the building as a whole, to their original configuration, strength, 

stiffness and deformation capacity. It does not include routine correction of non-conforming 

conditions resulting from the original construction or upgrades intended to result in improvement 

in future seismic performance of the building. Repair must typically be performed under a 

building permit, requiring submittal, to the building department of construction documents, 

inspection and testing reports. 

Sections 6.2 through 6.3 provide recommended methods of repair for various types of 

damage. These recommendations are not intended to be used for the routine repair of 

construction non-conformance commonly encountered in fabrication and erection work. Industry 

standard practices are acceptable for such repairs. Recommendations for assessment of the 

seismic performance capability of existing buildings and upgrade of buildings to improve 

performance capability may be found in a companion publication, FEMA-351, Recommended 

Seismic Evaluation and Upgrade Criteria for Existing Welded Steel Moment-Frame Buildings. 

Commentary: Based on the observed behavior of actual buildings in the 

Northridge earthquake, as well as recent test data, welded steel moment frame 

buildings constructed with the typical detailing and construction practice 

prevalent prior to 1994 do not have the deformation capacity they were presumed 

to possess at the time of their design and therefore present significantly higher 

risks than was originally thought. When these buildings are damaged or have 

latent construction defects, this risk is higher still. 

Based on limited testing, it is believed that the repair recommendations 

contained in these Recommended Criteria can be effective in restoring a 

building's pre-earthquake condition, and to the extent that the detailing, 

workmanship and materials of repair work are superior to the original 

construction, provide some marginal improvement in seismic performance 

capability. This does not imply, however, that the repaired building will be an 

acceptable seismic risk. As a minimum, it should be assumed that buildings that 

are repaired, but not upgraded, can sustain similar and possibly more severe 

damage in future earthquakes than they did in the present event. If this is 

unacceptable, either to the owner or the building official, then the building should 

be upgraded to provide improved future performance. Seismic upgrade can 

consist of local reinforcement of individual moment connections, column splices 

and other critical connections, as well as alteration of the basic lateral-forceresisting characteristics of the structure through addition, for example, of braced 

frames, shear walls, base isolation, and energy dissipation devices. Performance 

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FEMA-352 and Repair Criteria for Welded 

Chapter 6: Postearthquake Repair Steel Moment-Frame Buildings 

evaluation and structural upgrade are beyond the scope of these Recommended 

Criteria. Criteria for performance evaluation and structural upgrade may be 

found in a companion document, FEMA-351, Recommended Seismic Evaluation 

and Upgrade Criteria for Existing Welded Steel Moment-Frame Buildings. 

6.2 Shoring and Temporary Bracing 

6.2.1 Investigation 

Prior to engaging in repair activity, the structural engineer should investigate the entire 

building and perform an evaluation to determine if any imminent collapse or life-safety hazard 

conditions exist and to determine if the structure as a whole provides adequate stability to 

safeguard life during the repair process. The Level 2 evaluation process of Chapter 5 is one 

method of confirming both the building’s global structural stability, and the ability of individual 

connections to withstand ground shaking. Where hazardous conditions or lack of stability are 

detected, shoring and or temporary bracing should be provided prior to commencement of any 

repairs. 

Commentary: In projects relating to construction of new buildings, it is common 

practice to delegate all responsibility for temporary shoring and bracing of the 

structure to the contractor. Such practice may not be appropriate for severely 

damaged buildings. The structural engineer should work closely with the 

contractor to define shoring and bracing requirements. Some structural 

engineers may wish to perform the design of temporary bracing systems. If the 

contractor performs such design, the structural engineer should review the 

designs for adequacy and potential effects on the structure prior to 

implementation. 

6.2.2 Special Requirements 

Conditions that may become collapse or life-safety hazards during the repair operations 

should be considered in the development of repair details and specifications, whether they 

involve the damage area directly or indirectly. These conditions should be brought to the 

attention of the contractor by the structural engineer, and adequate means of shoring these 

conditions should be developed. Consideration should be given to sequencing of repair 

procedures for proper design of any required shoring. For column repair details that require 

removal of 20% or more of the damaged cross section, consideration should be given to the need 

for shoring to prevent overstress of elements due to redistribution of loads. 

Commentary: In general, contractors will not have adequate resources to define 

when such shoring is necessary. Therefore, the Contract Documents should 

clearly indicate when and where shoring is required. Design of this shoring may 

be provided by the structural engineer, or the contract documents may require 

that the contractor submit a shoring design, prepared by another registered 

structural engineer, to the structural engineer for review. 

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and Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings Chapter 6: Postearthquake Repair 

6.3 Repair Details 

The scope of repair work should be shown on drawings and specifications prepared by a 

structural engineer. The drawings should clearly indicate the areas requiring repair, as well as all 

repair procedures, details, and specifications necessary to properly implement the proposed 

repair. Sample repair details for various types of damage are included in these recommended 

criteria, for reference, only. 

Commentary: Examples of repair details are provided for some classes of 

damage, based on approaches successfully performed in the field following the 

1994 Northridge earthquake. Limited testing indicates these repair methods can 

be effective. Details are not complete in all respects and should not be used 

verbatim, as construction documents. Many repairs will require the application 

of more than one operation, as represented by a given detail. The sample details 

indicated may not be directly applicable to specific repair conditions. The 

structural engineer is cautioned to thoroughly review the conditions at each 

damaged element, connection or joint, and to determine the applicability and 

suitability of these details based on sound structural engineering judgment, prior 

to employing them on projects. 

In typical practice for construction of new buildings, the selection of means 

and methods used to construct design details are typically left to the contractor. 

In structural repair work, the members are typically under greater load and also 

restraint during the fabrication and erection process than is common in new 

construction. Therefore, the typical construction practices may not be 

appropriate and many contractors may not have the knowledge or experience to 

select appropriate methods for repair work. As a result, much greater 

specification of means and methods is recommended than is common in new 

construction. Although it is recommended that the engineer provide such 

specification as part of the construction documents, the engineer should also be 

open to suggestions for alternative procedures if the contractor desires to submit 

such procedures. If there is doubt as to the ability of alternative procedures to 

provide acceptable construction, a full-scale mock-up test of the proposed 

procedure should be considered. 

6.3.1 Approach 

Based on the nature and extent of damage several alternative approaches to repair should be 

considered. Repair approaches may include, but should not be limited to: 

• replacement of damaged portions of base metal (i.e. column and beam section), 

• replacement of damaged connection elements, 

• replacement of connection welds, or 

• repairs to portions of any of the aforementioned components. 

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Any or all of these techniques may be appropriate. The approaches used should consider 

adjacent structural components that may be affected by the repair or the effects of the repair. 

Where base material is to be removed and replaced with plates or shapes, clear direction 

should be given to orient the new material with the direction of rolling parallel to the direction of 

application of major axial loads to be resisted by the section. 

6.3.2 Weld Fractures - Type W Damage 

Prior to repair of fractures or rejectable defects in welds, sufficient material should be 

removed to completely eliminate any existing discontinuity or defect in the weld metal and if 

applicable, adjacent base material. Nondestructive Testing should be used to determine the 

extent of fracture or defect present and sufficient material should be removed to encompass the 

damaged area. It is suggested that material removal extend 2 inches beyond the apparent end of 

the fracture or defect. Simple fillet welds may be repaired by back gouging to eliminate unsound 

weld material and replacement of the damaged weld with sound material. Complete joint 

penetration (CJP) welds fractured through the full thickness should be replaced with sound 

material deposited in strict accordance with an appropriate Welding Procedure Specification 

(WPS) and the project specifications. Weld backing, existing end dams, and weld tabs should be 

removed from all welds that are being repaired. End dams should not be permitted in new work. 

After backing and tab elements are removed, the weld root should be back gouged to sound 

material, re-welded and a reinforcing fillet added. 

The structural engineer is cautioned to observe the provisions of AISC regarding intermixing 

of weld metals deposited by different weld processes (see AISC LRFD Manual of Steel 

Construction, second edition, page 6-77, and the 1989 AISC ASD Steel Construction Manual, 

ninth edition, page 5-69). As an example, E7018 shielded metal arc welding (SMAW) electrodes 

should not be used to weld over self-shielded flux cored arc welding (FCAW-S) deposits, unless 

appropriate precautions are taken (FEMA 355b). Typically, three to four passes of E7018 or 

similar notch tough filler metal should be deposited to ensure that the underlying FCAW-S filler 

metal has not degraded the overlying notch tough filler metal. Removed weld material from 

fractures not penetrating the full weld thickness should be replaced in the same manner as full 

thickness fractures. For other types of W damage, existing backing, end dams, and weld tabs 

should also be removed in a like manner to CJP weld replacement. Table 6-1 provides an index 

to suggested repair details for type W damage. 

Table 6-1 Reference Details for Type W Damage 

Damage or Defect Class Figure 

Rejectable defects at weld root Figure 6-1, Figure 6-2 

W2 Figure 6-3 

W3 Figure 6-3 

W4 Figure 6-3 

W5 Figure 6-3 

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20 oMin. 

Arc - Gouge 

1/4î radius min. 

Existing column flange 

Removed backing 

Existing beam flange 

Reweld & reinforce w/ fillet 20 o

Min. 

Arc - Gouge 

20 oMin. 

Arc - Gouge 

1/4î radius min. 

Existing column flange 

Removed backing 

Existing beam flange 

Reweld & reinforce w/ fillet 

Notes: 

1. Remove existing backing. 

2. Taper the depth of grinding or air arc gouging at each end to the face of flange with a minimum 2:1 

(horizontal/vertical) taper. 1⁄4". 

3. Grind all surfaces on which weld metal will be deposited. should be smooth, uniform and free from 

fins, tears, fractures and other discontinuities that would adversely affect weld strength. 

4. A fillet weld should be applied to reinforce the joint. 

the beam flange thickness, but not less than 1/4". need nor be more than 3/8". 

5. On joints to be repaired, remove all remaining weld tabs and excess weld metal beyond the length of the joint 

and grind smooth. Imperfection less than 1/16" should be removed by grinding. 

Provide a minimum root radius of 

Surfaces 

The size of the reinforcing fillet should be equal to 1/4 of 

It 

Repair as necessary. 

Figure 6-1 Gouge and Re-weld of Root Defect or Damage 

Existing column flange 

Removed backing 

Existing beam flange 

Backgouge, repair and 

reinforce 

Air-arc gouge Reweld 

Notes: 

1. Remove the entire fracture plus 1/8” of sound metal beyond each end. 

2. For additional notes, refer to Figure 6-1 

per Figure 6-1. 

Figure 6-2 Gouge and Re-weld of Fractured Weld 

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Existing column flange 

Existing beam flange 

Remove backing after

completing top welding,

Backgouge, repair and reinforce, per Figure 6-1. 

Air-arc gouge Reweld 

Existing column flange 

Existing beam flange 

Remove backing after

completing top welding,

Backgouge, repair and reinforce, per Figure 6-1. 

Remove backing after

completing top welding,

Backgouge, repair and reinforce, per Figure 6-1. 

Air-arc gouge Reweld 

For notes see Figure 6-1 and 6-2. 

Figure 6-3 Backgouge and Reweld Repair 

Commentary: Flux-cored arc-welding (FCAW-S) utilizes approximately 1-2% 

aluminum in the electrode to protect the weld from mixing with atmospheric 

nitrogen and oxygen. By itself, aluminum can reduce the toughness and ductility 

of weld metal. The design of FCAW-S electrodes requires the balance of other 

alloys in the deposit to compensate for the effects of aluminum. Other welding 

processes rely on fluxes and/or gasses to protect the weld metal from the 

atmosphere, relieving them of any requirement to contain aluminum or other 

elements that offset the effects of aluminum. If the original weld that is being 

repaired consists of FCAW-S and subsequent repair welds are made with 

shielded-metal arc-welding, SMAW (stick), using E7018, for example, the SMAW 

arc will penetrate into the FCAW-S deposit, resulting in the addition of some 

aluminum into the SMAW deposit. The notch toughness and/or ductility of the 

resultant weld metal may be substantially reduced as compared to pure E7018 

weld metal, based on the depth of penetration into the FCAW-S material. 

Various types of FCAW-S electrodes may be mixed one with the other without 

potentially harmful effect. Further, notch tough FCAW-S may be used to weld 

over other types of weld deposits without potentially harmful interaction. The 

structural engineer could specify all repairs on FCAW-S deposits be made with 

FCAW-S. Alternatively, intermixing of FCAW-S and other processes could be 

permitted provided the subsequent composition is demonstrated to meet material 

specification requirement, or adequate layers of reinforcing notch-tough filler 

metal are installed to avoid segregation (FEMA 355B). 

The recommendations contained in Chapters 3, 4, and 5 for inspection and 

evaluation of damaged buildings do not require extensive nondestructive 

examination of welds to detect defects or fractures that are not detectable by 

visual inspection but are rejectable under the AWS D1.1 provisions. 

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Nevertheless, it is likely that in the course of performing inspection and repair 

work, some such rejectable conditions will be found. It is recommended that any 

such detected conditions be repaired as part of the overall building repair 

program, as their presence in welds make the welds significantly more vulnerable 

to future fracturing under loading, particularly if the welds are composed of 

material with limited notch toughness. 

In the past, there has been considerable disagreement as to whether or not 

small cracks and defects at the root of a weld are earthquake damage or not. 

Proper observation by knowledgeable persons can reveal whether a root defect is 

a slag inclusion or lack of fusion, both conditions relating to the original 

construction, or an actual crack. It should be noted that cracks may not 

necessarily be caused by the building’s earthquake response. Some cracking 

invariably occurs in structures during the erection process as a result of residual 

stress conditions and thermal stresses. It is almost impossible to distinguish such 

cracks from those caused by an earthquake. Through detailed examination of the 

fracture surface for evidence of oxidation or other signs of age it may be possible 

to obtain clues as to when a crack initiated. Many researchers believe that the 

low toughness weld metal commonly used in construction prior to 1994 was 

incapable of arresting an earthquake induced fracture, once it initiated in a joint 

and that small cracks that do not penetrate through the metal are unlikely to be 

earthquake related. However, there have been reports from laboratory testing 

that indicate that small cracks do form in the weld metal and arrest prior to 

development of unstable fracture conditions, even in low-toughness weld metals. 

Therefore, without detailed examination of an individual fracture by 

knowledgeable individuals, no conclusive statement can be made as to whether 

weld cracking is earthquake induced. 

6.3.3 Column Fractures - Types C1 to C5 and P1 to P6 

Any column fracture observable with the naked eye or found by NDT and classified as 

rejectable in accordance with the AWS D1.1 criteria for Static Structures should be repaired. 

Repairs should include removing the fracture such that no sign of rejectable discontinuity or 

defect within a six (6) inch radius around the fracture remains. Removal should include 

eliminating any zones of fracture propagation, with a minimum of heat used in the removal 

process. Following removal of material, magnetic particle testing (MT) and/or Liquid Dye 

Penetrant testing (PT) should be used to confirm that all fractured material has been removed. 

Repairs of removed material may consist of replacement of portions of column section, build-up 

with weld material where small portions of column were removed, or local replacement of 

removed base metal with weld material. Procedures of weld fracture repair should be applied to 

limit the heat-affected area and to provide adequate ductility to the repaired joint. Table 6-2 

indicates representative details for these repairs. In many cases, it may be necessary to remove a 

portion of the girder framing to a column, in order to attain necessary access to perform repair 

work, per Figure 6-4. Refer to Section 6.3.5 and Figures 6-9 and 6-10 for repair of girders, or if 

access is restricted, as an alternative beam repair method. 

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Shore Beam 

Remove Shear Tab 

Replace upon completion 

Remove 

New web plate

thickness = tw + 1/8î 

tw 

Shore Beam 

Remove Shear Tab 

Replace upon completion 

Remove 

New web plate

thickness = tw + 1/8î 

tw 

portion of existing beam. 

Provide minimum 2î radius. 

Figure 6-4 Temporary Removal of Beam Section for Access 

When the size of divot (type C2) or transverse column fractures (types C1, C3, C4) dictate a 

total cut-out of a portion of a column flange or web (types P6, P7), the replacement material 

should be ultrasonically tested in accordance with ASTM A578-92, Straight-Beam Ultrasonic 

Examination of Plane and Clad Steel Plates for Special Applications, in conjunction with AWS 

K6.3 Shearwave Calibration. Acceptance criteria should be that of Level III. The replacement 

material should be aligned with the rolling direction matching that of the column. 

Table 6-2 Reference Details for Type C and P Damage 

Damage Class Figure 

Beam Access Figure 6-4 

C1 Figures 2-3, 6-4, 6-5 

C2 Figures 2-3, 6-4, 6-6 

C3 Figures 2-3, 6-4, 6-5 

C4 Figures 2-3, 6-4, 6-5 

C5 Figures 2-3, 6-4, 6-6 

P1 Figure 2-6; remove, prepare, replace 

P2 Figure 2-6; arc-gouge and reweld 

P4 Figure 2-6; arc-gouge and reweld 

P5 Figures 2-6, 6-7 

P6 Figures 2-6, 6-7 

P7 Figures 2-6, 6-7 

P8 Figures 2-6, 6-8 

Commentary: Special attention should be given to conditions where more than 

20% of the column cross section will be removed at one time, as special 

temporary shoring may be warranted. In addition, care should be taken when 

applying heat to a flange or web containing a fracture, as fractures have been 

observed to propagate with the application of heat. This can be prevented by 

drilling a small diameter hole at the end of the fracture, to prevent it from 

running. 

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Portion of Existing

beam flange removed Weld access hole 

in column web 

45 o 

per AWS D1.1

section 3.2.5,

and Figure 3.2 

Backgouge and reweld 

10o 

Notes: 

1. Investigate extent of fracture by UT to confirm that fractures are contained with the 45 degree angle zone 

of a standard pre-qualified CJP groove weld as defined in Figure 2.4, AWS D1.1, Joint Designation 

B-U4a-G 

2. Provide 10o bevel on lower flange plate, to channel slag out of joint. 

3. Grind all surfaces upon which weld metal will be deposited to smooth, uniform surface. 

Figure 6-5 Backgouge and Reweld of Column Flange 

New flange

splice plate 

Weld access 

hole and 

backing 

10 o 

6î minimum 

Note: Provide new flange plate material of the same strength, and width as the existing column flange. 

rolling direction of plate with that of column flange. 

existing flange with a tolerance of –0"/+1⁄4". 

flange to new flange plate welds prior to welding the column web to new flange plate. 

of the column flange, and upper edge of the splice plate down 10o, to channel slag out of joint. 

Align 

New plate should be of the same thickness as the 

The welding should be sequenced to connect the column 

Bevel the lower edge 

Figure 6-6 Replacement of Column Flange Repair 

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Column web Doubler Plate

Typical 

Web with Doubler Plate Web without Doubler Plate 

Notes: 

1. Prepare fractured section of doubler by air-arc gouging, grinding and rewelding, using web as backing. 

2. Prepare fractured section of web by air-arc gouging, grinding and rewelding, using doubler as backing or 

backgouge and reweld from reverse side, if no doubler present. 

Figure 6-7 Reweld Repair of Web plate and Doubler plate 

Flange removal and replacement 

per Figure 6-6, if required 

Weld access holes as required 

for weld terminations 

Notes: 

1. Sequence removal of portions of column and provide shoring as required to safely support existing column 

loads. 

2. Thickness of new web plate to match existing column web (tolerance –0"/+ "). 

Figure 6-8 Alternative Column Web Repair - Columns without Doubler Plates 

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6.3.4 Column Splice Fractures - Type C7 

Any fractures detected in column splices should be repaired by removing the fractured 

material and replacing it with sound weld material. For partial joint penetration groove welds, 

remove up to one half of the material thickness from one side and replace with sound material. 

Where complete joint penetration groove welds are required, it may be preferable to provide a 

double bevel weld, repairing one half of the material thickness completely prior to preparing and 

repairing the other half. Alternatively, if calculations indicate that column loads may safely be 

resisted with the entire section of column flange removed, or if suitable shoring is provided, it 

may be preferable to use a single bevel weld. 

Commentary: Special attention should be given to these conditions, as the 

removal of material may require special temporary shoring. Also, since partial 

penetration groove welds can serve as fracture initiators in tension applications, 

consideration should be given to replacing such damaged splice areas with 

complete joint penetration welds (see Figures 6-5 and 6-6). Also, the addition of 

flange plates to the outside face of each flange may be considered. 

6.3.5 Girder Flange Fractures - Type G3 to G5 

Repair of fractures in girder flanges may be performed by several methods. One method is to 

remove the fracture by air arc gouging such that no sign of discontinuity or defect within a six (6) 

inch radius around the fracture remains, preparing the surface by grinding and welding new 

material back. Alternatively, damaged portions of the girder flange may be removed and 

replaced with new plate as shown in Figure 6-9 or Figure 6-10. 

Commentary: Due to accessibility difficulties or excessive weld build-up 

requirements, it may become necessary to remove a portion of the girder flange to 

properly complete the joint repair. A minimum of six inches of girder flange may 

be removed to facilitate the joint repair, with the optimum length being equal to 

the flange width. After removal of the portion of flange, the face of column and 

cut edge of girder flange may then be prepared to receive a splice plate matching 

the flange in grade and width. Thickness should be adjusted as required to makeup the depth of the girder web and fillet removed as part of the preparation 

process. 

In the case of restricted access on one side of the beam (facade interference) it 

may be advantageous to make the plate narrower than the beam flange and 

perform all welding overhead. A CJP weld and fillet weld should be used to 

connect the plate to the column flange and beam flange, respectively. 

It is recommended that a double bevel joint be utilized in replacing the 

removed plate to eliminate the need for backing, consequently also eliminating 

the need for removal of the backing upon joint completion. A suggested joint 

detail is a B-U3/TC-U5, per AWS D1.1, with 1/3 tflange to 2/3 tflange bevels on the 

plate. The web of the girder should be prepared at the column and butt weld 

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areas to allow welding access. Weld tabs may be used at the column and beam 

flange butt welds, but should be removed following joint completion. The weld 

between the splice plate and the column flange should be completed first. If a 

double bevel weld is selected, the welder may choose to weld the first few passes 

from one face, then backgouge and weld from the second side. This may help to 

keep the interpass temperature below the maximum without down time often 

encountered in waiting for the weld to cool. 

Typ . 

New beam flange plate 

New web stiffeners, 

near side and far side 

Weld access 

hole 

Notes: 

1. New plate thickness to match beam flange thickness + height of removed web fillet. 

2. Weld sequence: a) weld of new flange plate to column; (b) weld of flange plate stiffeners to web and flange 

plate; (c) weld of new flange plate to beam flange; (d) weld of stiffener plate to beam flange and web. 

(

Figure 6-9 Beam Flange Plate Replacement 

New beam flange plate 

Figure 6-10 Alternative Beam Flange Plate Replacement 

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6.3.6 Buckled Girder Flanges - Type G1 

Where the top or bottom flange of a girder has buckled, and the rotation between the flange 

and web is less than or equal to the mill rolling tolerance given in the 1989 or 1994 AISC Manual 

of Steel Construction the flange need not be repaired. Where the angle is greater than mill rolling 

tolerance, repair should be performed and may consist of adding full height stiffener plates on the 

web over each portion of buckled flange, contacting the flange at the center of the buckle, (Figure 

6-11) or using heat straightening procedures. Another available approach is to remove the 

buckled portion of flange and replace it with plate, similar to Figures 6-9 and 6-10. 

New stiffener plates

each side, 

tplate= tweb 

Note: Provide stiffeners at beginning of buckle and at center of buckle 

Figure 6-11 Addition of Stiffeners at Buckled Girder Flange 

Commentary: Should flange buckling occur on only one side of the web, and the 

buckle repair consists of adding stiffener plates, only the side that has buckled 

need be stiffened. In case of partial flange replacement, special shoring 

requirements should be considered by the design engineer. 

6.3.7 Buckled Column Flanges - Type C6 

Any column flange or portion of a flange that has buckled to the point where it exceeds the 

rolling tolerances given in the AISC Manual of Steel Construction should be repaired. Flange 

repair may consist either of flame straightening or of removing the entire buckled portion of the 

flange and replacing it with material with yield properties similar to the actual yield properties of 

the damaged material similar to Figure 6-6. If workers with the appropriate skill to perform 

flame straightening are available, this is the preferred method. 

Commentary: For flange replacement, shoring is normally required. This 

shoring should be designed by the structural engineer, or may be designed by the 

contractor provided the design is reviewed by the structural engineer. 

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Flame straightening can be an extremely effective method of repairing 

buckled members. It is performed by applying heat to the member in a triangular 

pattern, in order to induce thermal strains that straighten the member out. Very 

large bends can be straightened by this technique. However, the practice of this 

technique is not routine and there are no standard specifications available for 

controlling the work. Consequently, the success of the technique is dependent on 

the availability of workers who have the appropriate training and experience to 

perform the work. During the heat application process, the damaged member is 

locally heated to very high temperatures. Consequently, shoring may be required 

for members being straightened in this manner. 

A number of references are available that provide more information on this 

process and its applications, published by AISC and others (Avent, 1992; 

Shonafelt and Horn, 1984) 

6.3.8 Gravity Connections 

Connections not part of the lateral force-resisting system may also be found to require repair 

due to excessive rotation or demand caused by distress of the lateral-force-resisting system. 

These connections should be repaired to a capacity at least equivalent to the pre-damaged 

connection capacity. Shear connections that are part of the lateral-force-resisting system should 

be repaired in a similar manner, with special consideration given to the nature and significance of 

the overall structural damage. 

Commentary: Testing of typical gravity shear connections conducted as part of 

the research performed in support of the development of these Recommended 

Criteria indicates that shear tab connections are capable of sustaining very large 

rotation demands without compromise of their gravity load carrying capacity 

(FEMA-355D). These connections tend to degrade in strength only when the 

imposed rotation becomes large enough to induce contact between the beam 

flange and the adjacent member. Once this contact occurs, rotational resistance 

of the connection stiffens substantially with large forces generated both through 

bearing of the beam flange against the adjacent member and as an axial force 

transmitted through the shear tab. The resulting forces can compromise, the 

bolts, the weld of the shear tab to the supporting member or the shear tab itself. 

Such behavior has not occurred in laboratory testing until connection rotations in 

excess of 0.1 radians were achieved. 

6.3.9 Reuse of Bolts 

Bolts in a connection displaying bolt damage or plate slippage should not be re-used, except 

as indicated herein. As indicated in the AISC Specification for Structural Joints using ASTM 

A325 or A490 Bolts (AISC, 1985), A490 bolts and galvanized A325 bolts should not be retightened and re-used under any circumstances. Other A325 bolts may be reused if determined to 

be in good condition. Touching up or re-tightening previously tightened bolts which may have 

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been loosened by the tightening of adjacent bolts need not be considered as reuse provided the 

snugging up continues from the initial position and does not require greater rotation, including 

the tolerance, than that required by Table 5 of the AISC Specification. 

Commentary: Proper performance of high strength bolts used in slip critical 

applications requires proper tensioning of the bolt. Although a number of 

methods are available to ensure that bolts are correctly tensioned, the most 

common methods relate to torquing of the nut on the bolt. When a bolt has been 

damaged, the torquing characteristics will be altered. As a result, damaged bolts 

may either be over-tightened or under-tightened, if reinstalled. The threads of 

ASTM A490 bolts and galvanized ASTM A325 bolts become slightly damaged 

when tightened, and consequently, should not be reused. To determine if an 

ungalvanized ASTM A325 bolt is suitable for re-use, a nut should be run up the 

threads of the bolt. If this can be done smoothly, without binding, then the bolt 

may be re-used. 

6.3.10 Welding Specifications 

Welded repairs involving thick plates and conditions of high restraint should be specified 

with caution. These conditions can lead to large residual stresses and in some cases, initiation of 

cracking before the structure is loaded. The potential for problems can be reduced by specifying 

appropriate joint configurations, welding processes, control of preheat, heat input during welding 

and cooldown, as well as selecting electrodes appropriate to the application. Engineers who do 

not have adequate knowledge to confidently specify these parameters should seek consultation 

from a person with the required expertise. 

6.4 Preparation 

6.4.1 Welding Procedure Specifications 

A separate Welding Procedure Specification (WPS) should be established for every different 

weld configuration, welding position, and material specification. The WPS is a set of focused 

instructions to the welders and inspectors stating how the welding is to be accomplished. Each 

type of weld should have its own WPS solely for the purpose of that weld. The WPS should 

include instructions for joint preparation based on material property and thickness, as well as 

welding parameters. Weld process, electrode type, diameter, stick-out, voltage, current, and 

interpass temperature should be clearly defined. In addition, joint preheat and postheat 

requirements should be specified as appropriate, including insulation guidelines if applicable. 

The WPS should also list any requirements that are mandated by the project specification. Two 

categories of qualified welding procedures are given in AWS D1.1. These are pre-qualified 

welding procedures and qualified-by-test welding procedures. Regardless of the type of 

qualification of a proposed welding procedure, a WPS should be prepared by the contractor and 

reviewed by the structural engineer responsible for the repairs. 

Commentary: Preparation of the WPS is normally the responsibility of the 

fabricator/erector. Sample formats for WPS preparation and submission are 

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included in AWS D1.1. Some contractors fill out the WPS by inserting references 

to the various AWS D1.1 tables rather than the actual data. This does not meet 

the intent of the WPS, which is to provide specific instructions to the welder and 

inspector on how the weld is to be performed. The actual values of the 

parameters to be used should be included in the WPS submittal. 

6.4.2 Welder Training 

Training of welders should take place at the outset of the repair operations. Welders and 

inspectors should be familiar with the WPS, and should be capable of demonstrating familiarity 

with each of its aspects. A copy of the WPS should be located on site, preferably at the 

connection under repair, accessible to all parties involved in the repair. 

6.4.3 Welder Qualifications 

Welders must be qualified and capable of successfully making the repair welds required. All 

welders should be qualified to the AWS D1.1 requirements for the particular welding process and 

position in which the welding is to be performed. Successful qualification to these requirements, 

however, does not automatically demonstrate a welder's ability to make repair welds for all the 

configurations that may be encountered. Specific additional training and/or experience may be 

required for repair situations. Inexperienced welders should demonstrate their ability to make 

proper repair welds. This may be done by welding on a mock-up assembly (see Section 6.4.4) 

that duplicates the types of conditions that would be encountered on the actual project. 

Alternatively, the welder could demonstrate proficient performance on the actual project, 

providing this performance is continuously monitored during the construction of at least the first 

weld repair. This observation should be made by a qualified welding inspector or engineer. 

6.4.4 Joint Mock-Ups 

A joint mock-up should be considered as a training and qualification tool for each type of 

repair that is more challenging than work in which the welder has previously demonstrated 

competence. This will allow the welder to become familiar with atypical welds, and will give the 

inspector the opportunity to observe clearly the performance of each welder. An entire mock-up 

is recommended for each such case, rather than only a single pass or portion of the weld. In a 

complete mock-up, all welding positions and types of weld would be experienced, thus showing 

the welder capable of both completing successfully the weld in all required positions, and 

applying all heating requirements. 

Commentary: The structural engineer may, at his or her discretion, require joint 

mock-ups to be performed for specific types of repair work as part of the project 

specifications. This practice is recommended where repair work must be made 

under conditions of unusual or restricted access, under conditions of high 

restraint, or for any joint that is not routinely performed in the industry. 

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6.4.5 Repair Sequence 

Repair sequence should be considered in the design of repairs, and any sequencing 

requirements should be clearly indicated on the drawings and WPS. Structural instabilities or 

high residual stresses could arise from improper sequencing. The order of repair of flanges, 

shear plates, and fractured columns should be indicated on the drawings as appropriate to guard 

against structural failure and to reduce possible residual stresses. 

6.4.6 Concurrent Work 

The maximum number of connections permitted to be repaired concurrently should be 

indicated on the drawings or in the project specifications. 

Commentary: Although a connection is damaged, it may still possess significant 

ability to participate in the structure's lateral-force-resisting system. 

Consideration should be given to limiting the total number of connections being 

repaired at any one time, as the overall lateral-force resistance of the structure 

may be temporarily reduced by some repair operations. If many connections are 

under repair simultaneously, the overall lateral resistance of the remaining frame 

connections may not be adequate to protect the structure's stability. Although this 

appears to fall under the category of means and methods, the typical contractor 

would have no way of determining the maximum number of connections that can 

be repaired at any one time without requiring supplemental lateral bracing of the 

building during construction. Therefore, the structural engineer should take a 

proactive role in determining this. 

6.5 Execution 

6.5.1 General 

FEMA-353, Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications, provides recommended general 

requirements to be included in specifications for repair. The following are of particular 

importance: 

• Strict enforcement of the welding requirements in AWS D1.1 adopted and modified by the 

building code. 

• Implementation of the special inspection requirements in the 1997 NEHRP Recommended 

Provisions for New Buildings and AWS D1.1, as well as such other requirements enforced by 

the local Building Department. Visual inspection means that the inspector inspects the 

welding periodically for adherence to the approved Welding Procedure Specification (WPS) 

and AWS D1.1, starting with preliminary tack welding and fit-up and proceeding through the 

welding process. Reliance on the use of nondestructive testing (NDT) at the end of the 

welding process alone should be avoided. Use visual inspection in conjunction with NDT to 

improve the chances of achieving a sound weld. 

• Require the fabricator to prepare and submit a WPS with at least the information required by 

AWS D1.1, as discussed in Section 6.5.4. 

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• Welding electrodes for welded joints in severe service and with significant consequences of 

failures should be capable of depositing weld metal with a minimum notch toughness of 40 

ft-lbs at the anticipated service temperature and 20 ft-lbs at 0oF. Refer to FEMA-353. Joints 

in this category include all complete joint penetration groove welds in beam and column 

flanges and webs, welds of continuity plates to column flanges and similar elements subject 

to large cyclic stresses at or near plastic levels. 

• All welds for the frame girder-column joints should be started and ended on weld run-off tabs 

where practical. All weld tabs should be removed, the affected area ground smooth and 

tested for defects using the magnetic particle method. Acceptance criteria should be per AWS 

D1.1, Section 8.15.1. Surface imperfections less than 1/16 inch in dimension should be 

removed by grinding. Deeper gouges, areas of lack of fusion, and slag inclusions, for 

example, should be removed by gouging or grinding and rewelding following the procedures 

outlined above. 

• Weld tabs should conform to the requirements of AWS D1.1. End dams should not be 

permitted. 

• Steel backing (backing bars), if used, should be removed from new and/or repaired welds at 

the girder bottom flange, the weld root back-gouged by air arcing and the area tested for 

defects using the magnetic particle method, as described above. The weld should be 

completed and reinforced with a fillet weld. Removal of the weld backing at repairs of the 

top girder flange weld may be considered, at the discretion of the structural engineer. 

6.5.2 Removal of Backing 

Prior to removing weld backing on existing joints, the contractor should prepare and submit a 

written WPS for review by the structural engineer. The WPS should conform to the 

requirements of AWS D1.1. In addition the contractor should propose the method(s) that will be 

used to remove the weld backing, back gouge to sound metal and when during this process 

preheat will be applied. 

Although project conditions may vary, steel backing may be removed either by grinding or by 

the use of air arc, or oxy-fuel gouging. The zone just beyond the theoretical 90 degree 

intersection of the beam to column flange should be removed by either air arc or oxy-fuel 

gouging followed by a thin grinding disk, or by a grinding disk alone. This shallow gouged 

depth of weld and base metal should then be tested by magnetic particle testing (MT) to 

determine if any linear indications remain. If the area is free of indications the area may then be 

re-welded. Preheat should be maintained and monitored throughout the process. If no further 

modification is to be made or if the modification will not be affected by a reinforcing fillet weld, 

the reinforcing fillet may be welded while the connection remains at or above the minimum 

preheat temperature and below the maximum interpass temperature. 

Commentary: Only removal of backing from the bottom-beam-flange-to-column 

joint is recommended. Removal of the weld backing from the top flange may be 

difficult, particularly along perimeter frames where access to the outer side is 

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restricted. Tests conducted to date have not been conclusive with regard to the 

benefit of top flange weld backing removal, and therefore, this is not generally 

recommended. There is no direct evidence that removal of weld backing from 

continuity plates in the column panel zone is necessary. 

6.5.3 Removal of Weld Tabs 

If weld tabs were used and are to be removed in conjunction with the removal of the weld 

backing, the tabs should be removed after the weld backing has been removed and fillet added. 

Weld tabs may be removed by air arc or oxy-fuel gouging followed by grinding or by grinding 

alone. The resulting contour should blend smoothly with the face of the column flange and the 

edge of the beam flange and should have a radius of 1/4-3/8 inch. The finished surface should be 

visually inspected for contour and any visually apparent indications. This should be followed by 

MT. Linear indications found in this location of the weld may be detrimental. They may be the 

result of the final residue of defects commonly found in the weld tab area. Linear indications 

should be removed by lightly grinding or using a cutting tool until the indication is removed. If 

after removal of the defect the ground area can be tapered and is not beyond the theoretical 90 

degree intersection of the beam flange edge and column flange, weld repair may not be necessary 

and should be avoided if possible. 

Existing end dams, if present, should be removed from joints undergoing repair. Prior to 

removal of end dams, the contractor should submit a removal / repair plan which lists the method 

of dam removal, defect removal, and welding procedure including, process, preheat, and joint 

configuration. The tab may be removed by grinding, air arc or oxy-fuel torch. 

6.5.4 Defect Removal 

Any rejectable weld defects should be removed by grinding or cutting tools, or by air arc 

gouging followed by grinding. The individual performing defect removal should be furnished 

the ultrasonic testing (UT) results which describe the location depth and extent of the defect(s). 

If defect removal extends into the theoretical weld section, weld repair may be necessary. 

The weld repair should be performed in accordance with the contractor's WPS, with strict 

adherence to the preheat requirements. The surface should receive a final visual inspection and 

MT after all repairs and surface conditioning has been completed. 

When the individual removing the defects has completed this operation, and has visually 

confirmed that no remnants remain, the surface should be tested by MT. Additional defect 

removal and MT may occur until the MT tests reveal that the defects have been removed. 

The contour of the surface at this point may be too irregular in profile to allow welding to 

begin. The surface should be conditioned by grinding or using a cutting tool to develop a joint 

profile that conforms to the WPS. Prior to welding, MT should be performed to determine if any 

additional defects have been exposed. 

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Based upon a satisfactory MT the joint may be prepared for welding. Weld tabs (and backing 

if necessary) should be added. The welding may begin and proceed in accordance with the WPS. 

The theoretical weld must be completed for its full height and length. Careful attention should 

be paid to ensure that individual weld bead size does not exceed that permitted by the WPS. 

If specified, the weld tabs and backing should be removed in accordance with the guideline 

section describing this technique. The final weld should be inspected by MT and UT. 

6.5.5 Girder Repair 

If at bottom flange repairs back gouging removes sufficient material such that a weld backing 

is required for the repair, after welding the backing should be removed from the girder. 

Alternatively, a double-beveled joint may be used The weld root should be inspected and tested 

for imperfections, which if found, should be removed by back-gouging to sound material. A 

reinforcing fillet weld should be placed at “T” joints. The reinforcing fillet should have a size 

equal to one-quarter of the girder flange thickness. It need not exceed 3/8 inch (see Note J, 

Figure 3.4 of AWS D1.1.) 

If the bottom flange weld requires repair, the following procedure may be considered: 

1. The root pass should not exceed a 1/4 inch bead size. 

2. The first half-length root pass should be made with one of the following techniques, at the 

option of the contractor: 

(a) The root pass may be initiated near the center of the joint. If this approach is used, the 

welder should extend the electrode through the weld access hole, approximately 1 inch 

beyond the opposite side of the girder web. This is to allow adequate access for clearing 

and inspection of the initiation point of the weld before the second half-length of the root 

pass is applied. It is not desirable to initiate the arc in the exact center of the girder width 

since this will limit access to the start of the weld during post-weld operations. After the 

arc is initiated, travel should progress towards the end of the joint (outboard beam flange 

edge), and the weld should be terminated on a weld tab. 

(b) The weld may be initiated on the weld tab, with travel progressing toward the center of 

the girder flange width. When this approach is used, the welder should stop the weld 

approximately 1 inch before the beam web. It is not advisable to leave the weld crater 

directly in the center of the beam flange width since this will hinder post-weld operations. 

3. The half-length root pass should be thoroughly slagged and cleaned. 

4. The end of the half-length root pass that is in the vicinity of the center of the beam flange 

should be visually inspected to ensure fusion, soundness, freedom from slag inclusions and 

excessive porosity. The resulting bead profile should be suitable for obtaining fusion by the 

subsequent pass to be initiated on the opposite side of the girder web. If the profile is not 

conducive to good fusion, the start of the first root pass should be ground, gouged, chipped or 

otherwise prepared to ensure adequate fusion. 

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5. The second half of the weld joint should have the root pass applied before any other weld 

passes are performed. The arc should be initiated at the end of the half-length root pass that 

is near the center of the beam flange, and travel should progress to the outboard end of the 

joint, terminating on the weld tab. 

6. Each weld layer should be completed on both sides of the joint before a new layer is 

deposited. 

7. Weld tabs should be removed and ground flush to the beam flange. Imperfections less than 

1/16 inch should be removed by grinding. Deeper gouges, areas of lack of fusion, and slag 

inclusions, for example, should be removed by gouging or grinding and rewelding following 

the procedures outlined above. 

6.5.6 Weld Repair (Types W2, or W3 and Defects) 

When W2, or W3 cracks are found, the column base metal should be evaluated using UT to 

determine if fractures have progressed into the flange. This testing should be performed both 

during the period of discovery and during repair. Similar procedures should be followed when 

making repairs to defects at weld roots. 

When a linear planar-type defect such as a crack or lack of fusion can be determined to 

extend beyond one-half the thickness of the beam flange, it is generally preferred to use a doublesided weld for repair (even though the fracture may not extend all the way to the opposite 

surface.) This is because the net volume of material that needs to be removed and restored is 

generally less when a double-sided joint is utilized. It also results in a better distribution of 

residual stresses since they are roughly balanced on either side of the center of the flange 

thickness. 

Repair of cracks and defects in welds may warrant total removal of the original weld, 

particularly if multiple cracks are present. If the entire weld plus some base metal is removed, 

care must be taken not to exceed the root opening and bevel limits of AWS D1.1 unless a 

qualified by test WPS is used. If this cannot be avoided one of two options is available: 

1. The beveled face of the beam and/or the column face may be built up (buttered) until the 

desired root opening and angle is obtained. 

2. A section of the flange may be removed and a splice plate inserted. 

Commentary: Building up base metal with welding is a less intrusive technique 

than removing large sections of the base metal and replacing with new plate. 

However, this technique should not be used if the length of build-up exceeds the 

thickness of the plate. 

6.5.7 Weld Overlays 

An alternative approach that can be considered when repairing or reinforcing pre-Northridge 

beam-flange to column-flange welded connections is to apply a weld overlay. This procedure is 

suitable for repairing beam-flange welds that have been classified as having rejectable weld 

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defects, as opposed to fractures. This application consists of encapsulating the existing low 

toughness weld metal with material with a higher notch toughness. The high notch toughness 

weld metal overlay is very resistant to fracture initiation from surface discontinuities within the 

range of defects that would go undetected, effectively converts surface defects and small 

fractures in the existing CJP weld into internal defects, and further, provides local reinforcement 

of the joint for resistance of applied stresses. This repair approach was developed by an 

independent task group of engineers and researchers in the Los Angeles area, following the 1994 

Northridge earthquake. A schematic arrangement of a weld overlay is shown in Figure 6-12. 

Nomenclature: 

Fyb = Minimum specified yield stress of beam (force / unit area)

Fyc = Minimum specified yield stress of column (force / unit area)

V = Total Shear acting on connection based upon nominal beam flexural strength at the

plastic hinge location (i.e., y 1. R M p +V , where Lc = clear distance between 

L gravity 

c 

column flanges) 

Ry = Ratio of Expected Yield Strength Fyc to the minimum specified yield strength Fyb 

SF = Total shear applied to each flange. (force) 

TF = Total tension applied to each flange. (force) 

h = Height of flaw assumed over full width of flange. (dimension) 

Figure 6-12 Weld Overlay Repair of Beam Flange to Column Flange Joint 

The design of the weld overlay is based on the premise that, in addition to transferring the 

flange flexural force to the column, the beam shear force must also be transferred through the 

beam flange weld. Based on physical testing of typical connections, supplemented by finite 

element analysis the stress distribution across the beam-flange is assumed to be parabolic as 

shown in Figure 6-13. Design procedures have been developed for two overlay conditions: 

Class A, which requires that the overlay take the full connection demand; and Class C, which 

assumes that the remaining existing weld has 50% of its original design capacity remaining. The 

throat thickness of the weld overlay, which may be regarded as an elongated fillet weld, is 

determined from geometric considerations similar to standard fillet welds. 

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Figure 6-13 Plan View and Assumed Stress Distribution for Weld Overlay Design 

Commentary: Independent research into the performance of weld overlay repairs 

indicates that beam-column connections repaired using this technique are 

substantially more rugged than typical unreinforced connections employing low 

toughness weld metals. Refer to Anderson, et al. (2000) for more detailed 

information on this technique. 

6.5.8 Column Flange Repairs - Type C2 

Damage type C2 is a pullout type failure of the column flange material. The failure surface 

should be conditioned to a concave surface by grinding and inspected for soundness using MT. 

The concave area may then be built up by welding (buttering). The joint contour described in the 

WPS should specify a "boat shaped" section with a "U" shaped cross section and tapered ends. 

The weld passes should be horizontal stringers placed in accordance with the WPS. Since 

stop/starts will occur in the finished weld, care must be taken to condition each stop/start to 

remove discontinuities and provide an adequate contour for subsequent passes. The final surface 

should be ground smooth and flush with the column face. This surface and immediate 

surrounding area should be subjected to magnetic particle testing and ultrasonic testing. 

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Steel Moment-Frame Buildings Appendix A: Detailed Procedures for Performance Evaluation 

APPENDIX A: DETAILED PROCEDURES FOR PERFORMANCE 

EVALUATION 

A.1 Scope 

This appendix provides detailed procedures for evaluating confidence levels associated with 

the ability of damaged WSMF structures to resist collapse in levels of ground shaking likely to 

occur in the period immediately following a major earthquake. These detailed procedures are 

provided as a supplement to the level 2 evaluation procedures in Chapter 5. They may be used to 

demonstrate enhanced levels of confidence with regard to the ability of a particular damaged 

structure to resist collapse, relative to the confidence levels that may be derived using the more 

simplified procedures of Chapter 5. The procedures of this appendix are required as a 

supplement to the Chapter 5 procedures for structures that are classified as irregular, considering 

the effects of the damage sustained. 

Commentary: Chapter 5 provides the basic procedures for a level 2 evaluation, 

using factored demand-to-capacity ratios to indicate a level of confidence with 

regard to a damaged building’s ability to resist collapse in that level of ground 

shaking likely to occur in the period immediately following a damaging 

earthquake. The tabular values of demand and resistance factors and confidence 

indices contained in Chapter 5 were derived using the procedures presented in 

this appendix, applied to the performance evaluation of a suite of model 

buildings. Since this suite of model buildings is not completely representative of 

any individual structure, the use of the tabular values inherently entails some 

uncertainty, and thus reduced levels of confidence, with regard to performance 

prediction. The detailed procedures in this appendix permit reduction in these 

uncertainties, and therefore enhanced confidence with regard to prediction of 

building performance. These more detailed procedures must be used for those 

building configurations, that is, irregular structures not well represented by the 

model buildings used as the basis for the values contained in Chapter 5. 

A.2 Performance Evaluation Approach 

A.2.1 Confidence of Ability to Withstand Collapse 

The evaluation procedures contained herein permit estimation of a level of confidence 

associated with the ability of a damaged building to withstand collapse for the levels of ground 

shaking likely to be experienced within a year following a damaging earthquake. 

Commentary: The probability that a damaged building may experience collapse 

as a result of ground shaking likely to occur in the year following a damaging 

event is calculated as the integral over a year’s time of the probability that 

damage will exceed the collapse capacity of the structure. Mathematically, this 

may be expressed as: 

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Appendix A: Detailed Procedures for Performance Evaluation Steel Moment-Frame Buildings 

P(D > PL) = � PD>PL (x)h(x)dx (A-1) 

where: 

P(D>PL) = Probability of damage exceeding the collapse level in a period 

of “t” years, taken as 1 year 

PD>PL(x) = Probability of damage exceeding a collapse level given that the 

ground motion intensity is level x, as a function of x 

h(x)dx = probability of experiencing a ground motion intensity of level 

(x) to (x + dx) in a period of 1 year following the first 

damaging event 

Vulnerability may be thought of as the capacity of the structure to resist collapse. 

Structural response parameters that may be used to measure capacity include the 

structure’s ability to undergo global building drift, maximum tolerable member 

forces and inelastic deformations. Ground accelerations associated with the 

seismic hazard, and the resulting global building drift, member forces and 

inelastic deformations produced by the hazard may be thought of as demands. If 

both the demand that a structure will experience over a period of time and the 

structure’s capacity to resist this demand could be perfectly defined, then the 

probability that damage could exceed a collapse level within a period of time, 

could be ascertained with 100% confidence. However, the process of predicting 

the capacity of a structure to resist ground shaking demands as well as the 

process of predicting the severity of demands that will actually be experienced 

entail significant uncertainties. Confidence level is a measure of the extent of 

uncertainty inherent in this process. A level of 100% confidence may be 

expressed as perfect confidence. In reality, it is never possible to attain such 

confidence. Confidence levels on the order of 90 or 95% are considered high, 

while confidence levels less than 50% are considered low. 

Generally, uncertainty can be reduced, and confidence increased, by 

obtaining better knowledge or using better procedures. For example, enhanced 

understanding and reduced uncertainty with regard to the prediction of the effects 

of ground shaking on a structure can be obtained by using a more accurate 

analytical procedure to predict the structure’s response. Enhanced 

understanding of the capacity of a structure to resist ground shaking demands can 

be obtained by obtaining specific laboratory data on the physical properties of the 

materials of construction and on the damageability of individual beam-column 

connection assemblies. 

The evaluation procedures of Chapter 5 are based on the typical 

characteristics of standard buildings. Since they are based on the capacity 

characteristics of typical rather than specific structures, the procedures contained 

in Chapter 5 inherently incorporate significant uncertainty in the performance 

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Steel Moment-Frame Buildings Appendix A: Detailed Procedures for Performance Evaluation 

prediction process. As a result of this significant uncertainty, it is anticipated that 

the actual ability of a structure to achieve a given performance objective may be 

significantly better than would be indicated by those simple procedures. The 

more detailed procedures of this appendix may be used to improve the definition 

of the actual uncertainties incorporated in the prediction of performance for a 

specific structure and thereby to obtain better confidence with regard to the 

prediction of performance for an individual structure. 

As an example, using the procedures of Chapter 5, it may be found that for a 

specific structure, there is only a 30% level of confidence that the structure is 

capable of resisting collapse for the levels of ground shaking likely to be 

experienced in a period of a year following a damaging event. This rather low 

level of confidence may be more a function of the uncertainty inherent in the 

procedures used to estimate the probability of collapse than the actual inadequate 

capacity of the building to resist collapse. In such a case, it may be possible to 

use the procedures contained in this appendix to reduce the uncertainty inherent 

in the performance estimation and find that instead, there may be as much as an 

80 % or 90% level of confidence, in resisting collapse. The difference in such 

findings can mean the difference between deciding that a building must be 

vacated or that it can continue to be occupied. 

In both the procedures of this appendix and Chapter 5, the uncertainties 

associated with estimation of the intensity of ground motion have been neglected. 

These uncertainties can be high, on the order of those associated with structural 

performance. Thus, the confidence estimated using these procedures is really a 

confidence with regard to structural performance, given an assumed seismicity, 

dominated by a single event, consisting of a repeat of the original damaging 

event, within a period of a year following the initial damaging shock. It is 

believed that this assumed seismicity is conservative, but credible. 

A.2.2 Basic Procedure 

As indicated in Chapter 5, a demand and resistance factor design (DRFD) format is used to 

associate a level of confidence with the probability that a building will be able to resist collapse 

in the level of ground shaking anticipated in the year following a damaging earthquake. The 

basic approach is to determine a confidence parameter, l, which may then be used, with 

reference to Table 5-7, to determine the confidence level that exists with regard to performance 

estimation. The confidence parameter, l, is determined from the factored demand-to-capacity 

equation: 

l = gg aD (A-2) fC 

where: 

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Appendix A: Detailed Procedures for Performance Evaluation Steel Moment-Frame Buildings 

C = median estimate of the capacity of the structure. This estimate may be obtained either 

by reference to default values contained in the tables of Chapters 5, or by more 

rigorous direct calculation of capacity using the procedures of this appendix. 

D = calculated demand on the structure, obtained from a structural analysis. 

g = a demand variability factor that accounts for the variability inherent in the prediction 

of demand related to assumptions made in structural modeling and prediction of the 

character of ground shaking. 

ga = an analysis uncertainty factor that accounts for the bias and uncertainty associated 

with the specific analytical procedure used to estimate structural demand as a function 

of ground shaking intensity. 

f = a resistance factor that accounts for the uncertainty and variability inherent in the 

prediction of structural capacity as a function of ground shaking intensity 

l = a confidence index parameter from which a level of confidence can be obtained by 

reference to Table 5-7. 

Several structural response parameters are used to evaluate structural performance. The 

primary parameter used for this purpose is interstory drift. Interstory drift is an excellent 

parameter for judging the ability of a structure to resist P-D instability and collapse. It is also 

closely related to plastic rotation demand, or drift angle demand, on individual beam-column 

connection assemblies, and therefore a good predictor of the performance of beams, columns and 

connections. Other parameters used in these guidelines include column axial compression and 

column axial tension. In order to determine a level of confidence with regard to the ability of a 

building to resist collapse for the level of ground shaking likely to occur in the year immediately 

following an earthquake, the following steps are followed: 

1. A best estimate of the ground shaking intensity that caused the initial damage in the 

building is developed. This can be done by reference to instrumental recordings of ground 

motion at the building or nearby sites, the use of standard attenuation relations, or ground 

shaking contour maps. For the purpose of this evaluation, it is assumed that an event of 

similar intensity at the building site is likely to occur within a one year period. Ground 

shaking intensity should be characterized by a 5% damped elastic response spectrum. 

2. A mathematical structural model is developed to represent the damaged building 

structure. Note that since damage can result in unsymmetrical structural response, it may be 

necessary to develop several models. The model(s) are then subjected to a structural analysis, 

using any of the methods contained in Chapter 5. This analysis predicts the median estimates 

of maximum interstory drift demand, maximum column compressive demand, and maximum 

column-splice tensile demand, for the assumed repeat of the original damaging earthquake. 

3. Median estimates of structural capacity are determined. Median estimates of the 

interstory drift capacity of the moment-resisting connections and the building frame as a 

whole are determined, as are median estimates of column compressive capacity and columnsplice tensile capacity. Interstory drift capacity for the building frame, as a whole, may be 

estimated using the default values of Chapter 5 for regular structures, or the detailed 

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Steel Moment-Frame Buildings Appendix A: Detailed Procedures for Performance Evaluation 

procedures of Section A.6 of this appendix may be used. These detailed procedures are 

required for irregular structures or for regular structures that have been made irregular by the 

damage they have sustained. Interstory drift capacity for moment-resisting connections may 

be estimated using the default values of Chapter 5, for typical connection types, or direct 

laboratory data on beam-column connection assembly performance capability and the 

procedures of Section A.5 of this appendix may be used. Median estimates of column 

compressive capacity and column splice tensile capacity are made using the procedures of 

Chapter 5. 

4. A factored-demand-to-capacity ratio l is determined. For each of the performance 

parameters, i.e. interstory drift as related to global building frame performance, interstory 

drift as related to connection performance, column compression, and column-splice tension, 

Equation A-2 is independently applied to determine the value of the confidence parameter l. 

In each case, the calculated estimates of demand D and capacity C are determined using steps 

3, and 4 respectively. If the procedures of Chapter 5 are used to determine either demand or 

median capacity estimates, than the corresponding values of the demand factors g and 

resistance factors f should also be determined in accordance with the procedures of that 

chapter. If the procedures of this appendix are used to determine median estimates of 

demand or capacity, then the corresponding demand and resistance factors should be 

determined in accordance with the applicable procedures of this appendix. 

5. Evaluate confidence. The confidence obtained with regard to the ability of the structure to 

meet the performance objective is determined using the lowest of the l values determined in 

accordance with Step 4 above, back-calculated from the equation: 

l = e -bbUT ( KX -k bUT / 2) (A-3) 

where: 

b = a coefficient relating the incremental change in demand (drift, force, or deformation) 

to an incremental change in ground shaking intensity, at the hazard level of interest 

typically taken as having a value of 1.0, 

bUT = an uncertainty measure equal to the vector sum of the logarithmic standard deviation 

of the variation in demand and capacity, resulting from uncertainty, 

k = the slope of the hazard curve, in ln-ln coordinates, at the hazard level of interest, i.e., 

the ratio of incremental change in SaT1 to incremental change in annual probability of 

exceedance. This is taken as having a value of 5, representative of the assumed 

seismicity for the year following a damaging earthquake, 

KX = standard Gaussian variate associated with the probability x of not being exceeded, as 

a function of number of standard deviations above or below the mean found in 

standard probability tables. 

Table 5-7 provides a solution for this equation, for various values of the parameters l and 

bUT. 

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Appendix A: Detailed Procedures for Performance Evaluation Steel Moment-Frame Buildings 

The values of the parameter bUT used in Equation A-3 and Table 5-7 are used to account for 

the uncertainties inherent in the estimation of demands and capacities. Uncertainty enters the 

process through a variety of assumptions that are made in the performance evaluation process, 

including for example assumed values of damping, structural period, properties used in structural 

modeling, and strengths of materials. Assuming that the amount of uncertainty introduced by 

each of the assumptions can be characterized, the parameter bUT can be calculated using the 

equation: 

bUT = 

2 �i u i b (A-4) 

where bui are the standard deviations of the natural logarithms of the variation in demand or 

capacity resulting from each of these various sources of uncertainty. Sections A.5 and A.6 

indicate how to determine bui values associated with demand estimation, beam-column 

connection assembly behavior, and building global stability capacity prediction respectively. 

A.3 Determination of Hazard Parameters 

In order to implement these postearthquake evaluation procedures, it is necessary to obtain an 

estimate of the 5% damped, linear response spectrum for the original damaging earthquake, and 

to obtain from that response spectrum, an estimate of the spectral response acceleration, SaT1at 

the fundamental period of the damaged building. 

A.4 Determination of Demand Factors 

The demand variability factor, g, and analytical uncertainty factor, ga, are used to adjust the 

calculated interstory drift, column axial load and column splice tension demands to their mean 

values, considering the variability and uncertainty inherent in drift demand prediction and 

probable intensity of ground shaking during the year following the initial damaging earthquake. 

Variability in drift demand prediction is primarily a result of the fact that due to relatively 

subtle differences in acceleration records, a structure will respond somewhat differently to 

different ground motion records, even if they are well characterized by the same response 

spectrum. Since it is not possible to predict the exact acceleration record that a structure may 

experience, it is necessary to account for the probable variation in demands produced by all 

possible different records. This is accomplished by developing a nonlinear mathematical model 

of the structure, and running nonlinear response history analyses of the structure for a suite of 

ground motion records, all of which are scaled to match the 5% damped linear spectral response 

acceleration, SaT1, described in Section A.3. From these analyses, statistics are developed for the 

median value and standard deviation of the natural logarithm of the various demand parameters 

including maximum interstory drift, column axial load, and column-splice tension. These 

standard deviations of the natural logarithms of these response parameters are denoted bDR. 

Once the value of bDR has been determined, the demand variability factor, g, is calculated 

from the equation: 

A-6
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and Repair Criteria for Welded FEMA 352 

Steel Moment-Frame Buildings Appendix A: Detailed Procedures for Performance Evaluation 

k 2 bDR 

g = e 2b (A-5) 

where: 

k is the logarithmic slope of the hazard curve, taken as having a value of 5. 

b is a coefficient that represents the amount that demand increases as a function of 

hazard, and may normally be taken as having a value of 1.0 

Uncertainty in the prediction of demands is due to an inability to define accurately the value 

of such parameters as the yield strength of the material, the viscous damping of the structure, the 

effect of nonstructural components, the effect of foundation flexibility on overall structural 

response, and similar modeling issues. Although it is not feasibly practical to do so, it is 

theoretically possible to measure each of these quantities for a building and to model their effects 

exactly. Since it is not practical to do this, instead, we use likely values for each of these effects 

in the model, and account for the possible inaccuracies introduced by using these likely values, 

rather than real values. These inaccuracies are accounted for by developing a series of models to 

represent the structure, accounting for the likely distribution of these various parameters. Each 

of these models is used to run analyses with a single ground motion record, and statistics are 

developed for the effect of variation in these parameters on predicted demands. As with the 

variability due to ground motion, the standard deviation of the natural logarithms of the response 

parameters are calculated, and denoted by bDU. 

This parameter is used to calculate the analytical uncertainty factor, ga. 

In addition to uncertainty in demand prediction, the analytical uncertainty factor ga also 

accounts for inherent bias, that is systematic under- or over- prediction of demand, inherent in an 

analytical methodology. Bias is determined by using the analytical methodology, for example, 

elastic modal analysis, to predict demand for a suite of ground motions and then evaluating the 

ratio of the demand predicted by nonlinear time history analysis of the structure to that predicted 

by the methodology for the same ground motion. This may be represented mathematically as: 

C = 

demand predicted by nonlinear time history analysis (A-6) B demand predicted by analysis method 

where CB is the bias factor. The bias factor that is applicable to a specific structure is taken as 

the median value of CB calculated from a suite of ground motions. The variation in the bias 

factors obtained from this suite of ground motions is used as one of the components in the 

calculation of b DU . 

Once the median bias factor, CB and logarithmic standard deviation in demand prediction 

b have been determined, the analysis uncertainty factor ga is calculated from the equation: DU

k b DU 

2 

g a = CBe 2b (A-7) 

A-7
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Recommended Postearthquake Evaluation 

FEMA 352 and Repair Criteria for Welded 

Appendix A: Detailed Procedures for Performance Evaluation Steel Moment-Frame Buildings 

The analysis uncertainty factors presented in Chapter 5 were calculated using this approach 

as applied to a suite of typical buildings. In addition to the uncertainties calculated using this 

procedure, it was assumed that even the most sophisticated methods of nonlinear time history 

analysis entail some uncertainty relative to the actual behavior of a real structure. Additional 

uncertainty was associated with other analysis methods to account for effects of structural 

irregularity which were not adequately represented in the suite of model buildings used in the 

study. The value of the total logarithmic uncertainty, b DU , used as a basis for the analysis 

uncertainty factors presented in Chapter 5 are summarized in Table A-1. The bias factors CB 

used in Chapter 5 are summarized in Table A-2. It is recommended that these default values for 

b DU and CB be used for all buildings. 

Table A-1 Default Logarithmic Uncertainty bDU for Various Analytical Methods 

Analysis Procedure 

Linear Static Linear 

Dynamic 

Nonlinear 

Static 

Nonlinear 

Dynamic 

Type 1 Connections 

Low Rise (<4 stories) 0.22 0.16 0.17 0.15 

Mid Rise (4 – 12 stories) 0.29 0.23 0.23 0.20 

High Rise (> 12 stories) 0.25 0.29 0.27 0.25 

Type 2 Connections 

Low Rise (<4 stories) 0.23 0.25 0.18 0.15 

Mid Rise (4 – 12 stories) 0.30 0.33 0.21 0.20 

High Rise (> 12 stories) 0.36 0.31 0.33 0.25 

Table A-2 Default Bias Factors CB 

Analysis Procedure 

Linear Static Linear 

Dynamic 

Nonlinear 

Static 

Nonlinear 

Dynamic 

Type 1 Connections 

Low Rise (<4 stories) 0.65 0.80 0.85 1.00 

Mid Rise (4 – 12 stories) 0.85 1.15 0.95 1.00 

High Rise (> 12 stories) 1.0 1.0 0.85 1.00 

Type 2 Connections 

Low Rise (<4 stories) 0.90 1.20 1.25 1.00 

Mid Rise (4 – 12 stories) 1.00 1.30 1.35 1.00 

High Rise (> 12 stories) 0.70 1.20 1.30 1.00 

A-8
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and Repair Criteria for Welded FEMA 352 

Steel Moment-Frame Buildings Appendix A: Detailed Procedures for Performance Evaluation 

Commentary: Although it may be possible, for certain structures, to increase the 

confidence associated with a prediction of probable earthquake demands on the 

structure, through calculation of structure-specific analysis uncertainty factors, in 

general this is a very laborious process. It is recommended that the default 

values, contained in Chapter 5 be used for most structures. The procedures 

contained in this appendix are most useful for calculating capacities and capacity 

factors. 

A.5 Determination of Beam-Column Connection Assembly Capacities 

The probable behavior of beam-column connection assemblies at various demand levels can 

best be determined by full-scale laboratory testing. Such testing can provide indications of the 

probable physical behavior of such assemblies in buildings. Depending on the characteristics of 

the assembly being tested, meaningful behaviors may include the following: onset of local 

buckling of flanges; initiation of fractures in welds, base metal or bolts; a drop in the moment 

developed by the connection beyond predetermined levels; or complete failure, at which point 

the connection is no longer able to maintain attachment between the beam and column under the 

influence of gravity loads. If sufficient laboratory data are available, it should be possible to 

obtain statistics, including a median value and standard deviation, on the demand levels at which 

these various behaviors occur. 

In the past, most laboratories used plastic rotation as the demand parameter by which beamcolumn connection assembly behavior was judged. However, since plastic deformations may 

occur at a number of locations within a connection assembly, including within the beam itself, 

within the connection elements and within the column panel zone or column, many laboratories 

have measured and reported plastic rotation angles from testing in an inconsistent manner. 

Therefore, in these Recommended Criteria, total interstory drift angle is the preferred demand 

parameter for reporting laboratory data. This parameter is less subject to interpretation by 

various testing laboratories and also has the advantage that it is approximately equal to the 

interstory drift angle predicted by linear structural analyses. Refer to FEMA 350, Recommended 

Seismic Design Criteria for New Steel Moment-Frame Buildings for additional information on 

laboratory testing protocols and parameters for reporting test behavior. 

Median drift angle capacities C and resistance factors f for common connection types are 

presented in Chapter 5. These values were determined from cyclic tests of full-size connection 

assemblies using the testing protocols indicated below. The cyclic tests are used to determine the 

load-deformation hysteresis behavior of the system and the connection drift angle at which the 

following behaviors occur: 

1. degradation of moment-resisting capacity of the assembly to a value below the nominal 

moment-resisting capacity, 

2. initiation of fracture of bolts, welds, or base metal that results in significant strength 

degradation of the assembly, and 

3. complete failure of the connection, characterized by an inability of the connection to 

maintain integrity of the beam to column connection under gravity loading. 

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FEMA 352 and Repair Criteria for Welded 

Appendix A: Detailed Procedures for Performance Evaluation Steel Moment-Frame Buildings 

Based on this data, drift angle statistics are obtained for a state of incipient collapse, qu. The 

quantity qu, the ultimate capacity of the connection, which occurs at damage state 3, above, is 

used to evaluate the acceptability of local connection behavior. 

A.5.1 Connection Test Protocols 

Two connection test protocols have been developed under this project. The standard protocol 

is intended to represent the energy input and cyclic deformation characteristics experienced by 

connection assemblies in steel moment frames, which are subjected to strong ground shaking 

from large magnitude earthquakes, but which are not located within a few kilometers of the fault 

rupture. This protocol (Krawinkler et al., 1997) is similar to that contained in ATC-24 and 

consists of ramped cyclic loading starting with initial cycles of low energy input, within the 

elastic range of behavior of the assembly and progressing to increasing deformation of the beam 

tip until assembly failure occurs. However, unlike ATC-24, the protocol incorporates fewer 

cycles of large-displacement testing to balance more closely the energy input to the assembly, 

with that likely experienced by framing in a real building. The second protocol is intended to 

represent the demands experienced by connection assemblies in typical steel moment-frame 

buildings responding to near-fault ground motion, dominated by large velocity pulses. This 

protocol (Krawinkler, 2000) consists of an initial single large displacement, representing the 

initial response of a structure to a velocity pulse, followed by repeated cycles of lesser 

displacement. 

Performance characteristics of connection assemblies, for use in performance evaluation of 

buildings, should be selected based on the characteristics of earthquakes dominating the hazard 

for the building site, at the specific hazard level. Most buildings are not located on sites that are 

likely to be subjected to ground shaking with near-field pulse characteristics. Connection 

performance data for such buildings should be based on the standard protocols. Buildings on 

sites that are proximate to a major active fault are most likely to experience ground shaking with 

these strong pulse-like characteristics and connection performance for such buildings should be 

based on the near-fault protocol. 

A.5.2 Determination of Beam-Column Assembly Capacities and Resistance Factors 

Median drift angle capacities for the quantity qU, should be taken directly from available 

laboratory data. The median value should be taken as that value from all of the available tests 

that is not exceeded by 50% of the tests. The value of the quantity f should be determine by the 

following procedure. 

1. Obtain the logarithmic standard deviation of the qU values available from the laboratory data. 

That is, take the standard deviation of the natural logarithms of the qU values obtained from 

each laboratory test. Logarithmic standard deviation may be determined from the formula: 

b = 

( )

1 

ln ln 

1 

2 

-

� - =

n 

x x n 

i i i (A-8) 

where: 

A-10
Recommended Postearthquake Evaluation 

and Repair Criteria for Welded FEMA 352 

Steel Moment-Frame Buildings Appendix A: Detailed Procedures for Performance Evaluation 

b = the standard deviation of the natural logarithms of the test data

xi = individual test data value

n = the number of tests from which data is available

ln xi = the mean of the logarithms of the xi values, 

2. Calculate the connection resistance factor due to randomness, the observed variation in 

connection behavior from laboratory testing using the equation: 

2.5 2

fR = e - b (A-9) 

3. Determine the connection resistance factor accounting for random and uncertain behaviors 

from the equation: 

f = fUfR = 0.9fR (A-10) 

where: 

fR = the resistance factor accounting for random behavior 

fU = the resistance factor accounting for uncertainty in the relationship between laboratory 

findings and behavior in real buildings and assumed in these recommended criteria to 

have a logarithmic standard deviation bU of 0.2. 

A.6 Global Stability Capacity 

In addition to consideration of local behavior, that is, the damage sustained by individual 

beams and beam-column connection assemblies, it is also important to consider the global 

stability of the frame. The procedures indicated in this section are recommended for determining 

an interstory drift capacity C and resistance factor f, associated with global stability of the 

structure. 

The global stability limit is determined using the Incremental Dynamic Analysis (IDA) 

technique (Cornell, 1999). This requires the following steps: 

1. Choose a suite of ten to twenty accelerograms representative of the site and hazard level for 

which the Collapse Prevention level is desired to be achieved. 

2. Select one of these accelerograms and perform an elastic time-history analysis of the 

building. Determine a scaling factor for this accelerogram such that the elastic time history 

analysis would result in response that would produce incipient yielding in the structure. 

Determine the 5%-damped, spectral response acceleration SaT1 for this scaled accelerogram at 

the fundamental period of the structure. On a graph with an abscissa consisting of peak 

interstory drift and an ordinate axis of SaT1, plot the point consisting of the maximum 

calculated interstory drift from the scaled analysis and the scaled value of SaT1. Draw a 

A-11
Recommended Postearthquake Evaluation 

FEMA 352 and Repair Criteria for Welded 

Appendix A: Detailed Procedures for Performance Evaluation Steel Moment-Frame Buildings 

straight line from the origin of the axes to this point. The slope of this line is referred to as 

the elastic slope, Se. 

3. Increase the scaling of the accelerogram, such that it will produce mild non-linear behavior 

of the structure. Perform a nonlinear time history analysis of the building for this scaled 

accelerogram. Determine the SaT1 for this scaled accelerogram and the maximum predicted 

interstory drift from the analysis. Plot this point on the graph. Call this point D1. 

4. Increase the scaling amplitude of the accelerogram slightly and repeat Step 3. Plot this point 

as D2. Draw a straight line between points D1 and D2. 

5. Repeat Step 4 until the straight line slope between consecutive points Di and Di+1, is less than 

0.2 Se. When this condition is reached, Di+1 is the global drift capacity for this accelerogram. 

If Di+1 > 0.10 then the drift capacity is taken as 0.10. Figure A-1 presents a typical series of 

plots obtained from such analyses. 

6. Repeat Steps 2 through 5 for each of the accelerograms in the suite selected as representative 

of the site and hazard and determine an interstory drift capacity for the structure for each 

accelerogram. 

7. Determine a median interstory drift capacity for global collapse C as the median value of the 

calculated set of interstory drift capacities, determined for each of the accelerograms. Note 

that the median value is that value exceeded by 50% of the accelerograms. 

0 

0.5 

1 

1.5 

2 

2.5 

0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 

Max imum Interstory Drift 

Spectral Acceleration (at fundame ntal period), g 

LA23 

LA22 

LA30 

LA24 

LA28 

Figure A-1 Representative Incremental Dynamic Analysis Plots 

8. Determine a logarithmic standard deviation for random differences in ground motion 

accelerograms, bR, using Equation A-8 of Section A.5.2. In this equation, xi is the interstory 

A-12
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and Repair Criteria for Welded FEMA 352 

Steel Moment-Frame Buildings Appendix A: Detailed Procedures for Performance Evaluation 

drift capacities predicted for the i

th accelerogram, and n is the number of accelerograms 

contained in the analyzed suite. 

9. Calculate the global resistance factor fR due to randomness in the predicted global collapse 

capacity for various ground motions from the equation: 

k b 2 -

f R = e 2b (A-10) 

where k and b are the parameters described in Section A.5.2 and b is the logarithmic standard 

deviation calculated in the previous step. 

10. Determine a resistance factor for global collapse from the equation: 

- k bU 

2 

f = fU f R = e 2b fR (A-11) 

where: 

fR is the global resistance factor due to randomness determined in Step 9. 

bU is the logarithmic standard deviation related to uncertainty in the analytical prediction of 

global collapse prevention taken as having a value of 0.15 for low-rise structures, 3 

stories or less in height; a value of 0.2 for mid-rise structures, 4 stories to 12 stories in 

height; and taken as having a value of 0.25 for high-rise structures, greater than 12 stories 

in height. 

It is important that the analytical model used for determining the global drift demand be as 

accurate as possible. The model should include the elements of the steel moment frame as well 

as framing that is not intended to participate in lateral load resistance. A nominal viscous 

damping of 3% of critical is recommended for most structures. The element models for beamcolumn assemblies should realistically account for the effects of panel zone flexibility and 

yielding, element strain hardening and stiffness and strength degradation, so that the hysteretic 

behavior of the element models closely matches that obtained from laboratory testing of 

comparable assemblies. 

Commentary: As noted above, accurate representation of the hysteretic behavior 

of the beam-column assemblies is important. Earthquake-induced global collapse 

initiates when displacements produced by the response to ground shaking are 

large enough to allow P-D instabilities to develop. Prediction of the onset of P-D 

instability due to ground shaking is quite complex. It is possible that an 

acceleration record will displace a structure to a point where static P-D 

instability would initiate, only to bring the structure back again before collapse 

can occur, due to a reversal in ground shaking direction. 

The basic effect of P-D instability is that a negative stiffness is induced in the 

structure. That is, P-D effects produce a condition in which increased 

displacement can occur at reduced lateral force. A similar and equally 

dangerous effect can be produced by local hysteretic strength degradation of 

A-13
Recommended Postearthquake Evaluation 

FEMA 352 and Repair Criteria for Welded 

Appendix A: Detailed Procedures for Performance Evaluation Steel Moment-Frame Buildings 

beam-column assemblies (FEMA-355C). Hysteretic strength degradation 

typically occurs after the onset of significant local buckling in the beam-column 

assemblies. It is important when performing Incremental Dynamic Analyses 

(IDA) that these local strength degradation effects, which show up as a concave 

curvature in the hysteretic loops in laboratory data, are replicated by the 

analytical model. Nonlinear analysis software that is currently commercially 

available is not in general able to model this behavior. Increasing the amount of 

dead load on the structure, to produce artificially the appropriate negative 

stiffness, can account approximately for these effects. 

A-14
Recommended Postearthquake Evaluation 

and Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings Appendix B: Sample Placards 

APPENDIX B: SAMPLE PLACARDS

B.1 “Inspected” Placard 

Title: 

(Iplacard.eps) 

Creator: 

Adobe Illustrator(R) 8.0 

Preview: 

This EPS picture was not saved 

with a preview included in it. 

Comment: 

This EPS picture will print to a 

PostScript printer, but not to 

other types of printers. 

Notes: 

1. Recommended placard color: green. 

2. Use for posting conditions Green-1, Green-2, and Green-3 of Table 3-2. 

3. Placard is from ATC (1995) 

B-1
Recommended Postearthquake Evaluation 

FEMA-352 and Repair Criteria for Welded 

Appendix B: Sample Placards Steel Moment-Frame Buildings 

B.2 “Restricted Use” Placard 

Title: 

(Rplacard.eps) 

Creator: 

Adobe Illustrator(R) 8.0 

Preview: 

This EPS picture was not saved 

with a preview included in it. 

Comment: 

This EPS picture will print to a 

PostScript printer, but not to 

other types of printers. 

Notes: 

1. Recommended placard color: yellow. 

2. Use for posting conditions Yellow-1 and Yellow -2 of Table 3-2. 

3. Placard is from ATC (1995) 

B-2
Recommended Postearthquake Evaluation 

and Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings Appendix B: Sample Placards 

B.3 Modified “Restricted Use” Placard 

Title: 

(Mplacard.eps) 

Creator: 

Adobe Illustrator(R) 8.0 

Preview: 

This EPS picture was not saved 

with a preview included in it. 

Comment: 

This EPS picture will print to a 

PostScript printer, but not to 

other types of printers. 

Notes: 

1. Recommended placard color: yellow. 

2. Use for posting conditions Yellow-1 and Yellow -2 of Table 3-2. 

3. Placard is from ATC (1995) 

B-3
Recommended Postearthquake Evaluation 

FEMA-352 and Repair Criteria for Welded 

Appendix B: Sample Placards Steel Moment-Frame Buildings 

B.4 “Unsafe” Placard 

Title: 

(Uplacard.eps) 

Creator: 

Adobe Illustrator(R) 8.0 

Preview: 

This EPS picture was not saved 

with a preview included in it. 

Comment: 

This EPS picture will print to a 

PostScript printer, but not to 

other types of printers. 

Notes: 

4. Recommended placard color: red. 

5. Use for posting conditions Red-1 and Red –2 of Table 3-2. 

6. Placard is from ATC (1995) 

B-4
Recommended Postearthquake Evaluation 

and Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings Appendix C: Sample Inspection Forms 

APPENDIX C: SAMPLE INSPECTION

Figure C-1 Inspection Form – Major Axis Column Connection 

C-1
Recommended Postearthquake Evaluation 

FEMA-352 and Repair Criteria for Welded 

Appendix C: Sample Inspection Forms Steel Moment-Frame Buildings 

Figure C-2 Inspection Form – Large Discontinuities – Major Axis 

C-2
Recommended Postearthquake Evaluation 

and Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings Appendix C: Sample Inspection Forms 

Figure C-3 Inspection Form – Minor Axis Column Connection 

C-3
Recommended Postearthquake Evaluation 

FEMA-352 and Repair Criteria for Welded 

Appendix C: Sample Inspection Forms Steel Moment-Frame Buildings 

Figure C-4 Inspection Form – Large Discontinuities – Minor Axis 

C-4
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

REFERENCES, BIBLIOGRAPHY, AND ACRONYMS 

This section contains references, additional bibliography and acronyms that are generally 

common to the set of reports, FEMA-350, FEMA-351, FEMA-352, and FEMA-353. Following 

the regular references are three sections containing ASTM Standards published by the American 

Society for Testing and Materials, West Conshohocken, Pennsylvania and listed numerically, 

AWS Specifications published by the American Welding Society, Miami, Florida, and listed 

numerically, FEMA Reports published by the Federal Emergency Management Agency, 

Washington, DC, and listed by report number, and SAC Reports published by the SAC Joint 

Venture, Sacramento, California, and listed by report number. 

References and Additional Bibliography. 

AISC, 1985, Specification for Structural Joints using ASTM A325 or A490 Bolts, American 

Institute of Steel Construction, Chicago, Illinois. 

AISC, 1989, Manual of Steel Construction, ASD, Ninth Edition, American Institute of Steel 

Construction, Chicago, Illinois. 

AISC, 1993, 1997, Load and Resistance Factor Design Specifications for Structural Steel 

Buildings, American Institute of Steel Construction, Chicago, Illinois. 

AISC, 1994a, Proceedings of the AISC Special Task Committee on the Northridge Earthquake 

Meeting, American Institute of Steel Construction, Chicago, Illinois. 

AISC, 1994b, Northridge Steel Update 1, American Institute of Steel Construction, Chicago, 

Illinois. 

AISC, 1997, Seismic Provisions for Structural Steel Buildings, American Institute of Steel 

Construction, Chicago, Illinois. 

AISC, 1998a, Load and Resistance Factor Design Specifications for Structural Steel Buildings, 

American Institute of Steel Construction, Chicago, Illinois. 

AISC, 1998b, LRFD Manual of Steel Construction, 2nd Edition, American Institute of Steel 

Construction, Chicago, Illinois. 

AISC, 1999, Supplement No. 1 to the 1997 Seismic Provisions for Structural Steel Buildings, 

American Institute of Steel Construction, Chicago, Illinois. 

Allen, J., Partridge, J.E., Richard, R.M., and Radau, S., 1995, “Ductile Connection Designs for 

Welded Steel Moment Frames,” Proceedings, 64th Annual Convention, Structural Engineers 

Association of California, Sacramento, California. 

Anderson, J, Duan, J., Xiao, Y., and Maranian, P., 2000, Improvement of Welded Connections 

Using Fracture Tough Overlays, Report No. SAC/BD-00/20, SAC Joint Venture, 

Sacramento, California. 

ASCE, 1998, ASCE-7 maps, American Society of Civil Engineers, Reston, Virginia. 

ASTM citations: see the list of ASTM Standards on page R-4. 

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Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

ATC, 1985, Earthquake Damage Evaluation Data for California, Report ATC-13 , Applied 

Technology Council, Redwood City, California. 

ATC, 1987, Evaluating the Seismic Resistance of Existing Buildings, Report ATC-14, Applied 

Technology Council, Redwood City, California. 

ATC, 1989, Procedures for Postearthquake Safety Evaluations of Buildings, Report ATC-20, 

Applied Technology Council, Redwood City, California. 

ATC, 1992, Guidelines for Cyclic Seismic Testing of Components of Steel Structures, Report 

ATC-24, Applied Technology Council, Redwood City, California. 

ATC, 1995, Addendum to the ATC-20 Postearthquake Building Safety Evaluation Procedures, 

Report ATC-20-2, Applied Technology Council, Redwood City, California. 

ATC, 1997a, Seismic Evaluation and Retrofit of Concrete Buildings, prepared by the Applied 

Technology Council (Report No. ATC-40), for the California Seismic Safety Commission 

(Report No. SSC 96-01), Sacramento, California. 

ATC, 1997b, NEHRP Guidelines for the Seismic Rehabilitation of Buildings, Report No. FEMA273, prepared by the Applied Technology Council for the Building Seismic Safety Council, 

published by the Federal Emergency Management Agency, Washington, DC. 

ATC, 1997c, Commentary to NEHRP Guidelines for the Seismic Rehabilitation of Buildings, 

Report No. FEMA-274, prepared by the Applied Technology Council for the Building 

Seismic Safety Council, published by the Federal Emergency Management Agency, 

Washington, DC. 

Avent, R., 1992, “Designing Heat-Straightening Repairs,” National Steel Construction 

Conference Proceedings, Las Vegas, Nevada. 

AWS citations: see the list of AWS reports, specifications and codes on page R-5. 

Barsom, J.M., 1996, “Steel Properties — Effects of Constraint, Temperature, and Rate of 

Loading,” Proceedings of the 2nd US Seminar, Seismic Design, Evaluation and Retrofit of 

Steel Bridges, San Francisco, Report No. UCB/CEE STEEL-96/09, Dept. of Civil and 

Environmental Engineering, UC Berkeley, pp.115-143. 

Boore, D.M., and Joyner, W.B., 1994, Proceedings of Seminar on New Developments in 

Earthquake Ground-Motion Estimation and Implications for Engineering Design Practice, 

Report ATC-35-1, Applied Technology Council, Redwood City, California, pp 6-1 to 6-41. 

BSSC, 1992, NEHRP Handbook for the Seismic Evaluation of Existing Buildings, developed by 

the Building Seismic Safety Council for the Federal Emergency Management Agency, 

Report FEMA-178, Washington, D.C. 

BSSC, 1997a, 1997 NEHRP Recommended Provisions for Seismic Regulations for New 

Buildings and Other Structures, Part 1 – Provisions, prepared by the Building Seismic 

Safety Council for the Federal Emergency Management Agency (Report No. FEMA-302), 

Washington, DC. 

BSSC, 1997b, 1997 NEHRP Recommended Provisions for Seismic Regulations for New 

Buildings and Other Structures, Part 2 – Commentary, prepared by the Building Seismic 

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Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

Safety Council for the Federal Emergency Management Agency (Report No. FEMA-303), 

Washington, DC. 

Campbell, K.W., and Bozorgnia, Y., 1994, “Near-Source Attenuation of Peak Horizontal 

Acceleration from Worldwide Accelerograms Recorded from 1957 to 1993,” Fifth U.S. 

National Conference on Earthquake Engineering, Proceedings, Vol. III, pp 283-292, 

Earthquake Engineering Research Institute, Oakland, California. 

Chi, W.M., Deierlein, G., and Ingraffea, A., 1997, “Finite Element Fracture Mechanics 

Investigation of Welded Beam-Column Connections,” SAC Joint Venture, Report No. 

SAC/BD-97/05. 

FEMA citations: see the list of FEMA reports on page R-6. 

Goel, R.K., and Chopra, A.K., 1997, “Period Formulas for Moment-Resisting Frame Buildings,” 

Journal of Structural Engineering, Vol. 123, No. 11, pp. 1454-1461. 

Gross, J.L., Engelhardt, M.D., Uang, C.M., Kasai, K. and Iwankiw, N.R., 1999, Modification of 

Existing Welded Steel Moment Frame Connections for Seismic Resistance, AISC Design 

Guide Series 12, American Institute of Steel Construction, Chicago, Illinois. 

Grubbs, K., 1997, “The Effect of the Dogbone Connection on the Elastic Stiffness of Steel 

Moment Frames” Masters Thesis, Department of Civil Engineering, University of Texas at 

Austin. 

ICBO, 1988, 1991, and 1997, Uniform Building Code, indicated edition, International 

Conference of Building Officials, Whittier, California. 

ICC, 2000, International Building Code, International Code Council, Falls Church, Virginia. 

Kircher, C.A., Nassar, A.A., Kustu, O. and Holmes, W.T., 1997, “Development of Building 

Damage Functions for Earthquake Loss Estimation,” Earthquake Spectra, Vol. 13, No. 4, 

Earthquake Engineering Research Institute, Oakland, California, pp. 663-682. 

Kircher, C.A., Reitherman, R.K., Whitman, R.V., and Arnold, C., 1997, “Estimation of 

Earthquake Losses to Buildings,” Earthquake Spectra, Vol. 13, No. 4, Earthquake 

Engineering Research Institute, Oakland, California, pp. 703-720. 

Kircher, C.A., 1999, Procedures for Development of HAZUS-Compatible Building-Specific 

Damage and Loss Functions, National Institute of Building Sciences, Washington, D.C. 

Krawinkler, H., Gupta, A., Medina, R. and Luco, N., 2000, Loading Histories for Seismic 

Performance Testing of SMRF Components and Assemblies, Report No. SAC/BD-00/10, 

SAC Joint Venture, Sacramento, California. 

NIBS, 1997a, HAZUS Earthquake Loss Estimation Methodology, Users Manual, National 

Institute of Building Sciences, Washington, DC. 

NIBS, 1997b, HAZUS Earthquake Loss Estimation Methodology, Technical Manual, 3 Volumes. 

National Institute of Building Sciences, Washington, DC. 

RCSC, 1996, Load and Resistance Factor Design: Specification for Structural Joints Using 

ASTM A325 or A490 Bolts, Research Council on Structural Connections. 

R-3
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

Richard, R., Partridge, J.E., Allen, J., and Radau, S., 1995, “Finite Element Analysis and Tests of 

Beam-to-Column Connections,” Modern Steel Construction, Vol. 35, No. 10, pp. 44-47, 

American Institute of Steel Construction, Chicago, Illinois. 

SAC citations: see the list of SAC Joint Venture reports on page R-7. 

Shonafelt, G.O., and Horn, W.B, 1984, Guidelines for Evaluation and Repair of Damaged Steel 

Bridge Members, NCHRP Report 271, prepared by the National Cooperative Highway 

Research Program, for the Transportation Research Board, Washington, DC. 

Wald, D.J., Quitoriano, T.H., Kanamori, H. and Scrivner, C.W., 1998, “Trinet Shakemaps – 

Rapid Generation of Peak Ground Motion and Intensity Maps for Earthquakes in Southern 

California”, SMIP98 Proceedings, California Division of Mines and Geology, Sacramento, 

California. 

Whitman, R., Anagnos, T., Kircher, C., Lagorio, H.J., Lawson, R.S., and Schneider, P., 1997, 

“Development of a National Earthquake Loss-Estimation Methodology,” Earthquake 

Spectra, Vol. 13, No. 4, Earthquake Engineering Research Institute, Oakland, California, 

pp. 643-661. 

Youssef, N.F.G, Bonowitz, D., and Gross, J.L., 1995, A Survey of Steel Moment-Resisting Frame 

Buildings Affected by the 1994 Northridge Earthquake, Report No. NISTR 56254, National 

Institute for Science and Technology, Gaithersburg, Maryland. 

ASTM Standards. 

ASTM Standards are published by the American Society for Testing and Materials, West 

Conshohocken, Pennsylvania, and are listed alphanumerically. 

ASTM, 1997, Standard Test Methods and Definitions for Mechanical Testing of Steel Products 

A6, Supplementary Requirement S5 

A36, Specification for Carbon Structural Steel 

A325, Specification for Structural Bolts, Steel, Heat-Treated, 120/105 ksi Minimum Tensile 

Strength 

A435, Straight Beam Ultrasonic Examination of Steel Plates 

A490, Specification for Heat-Treated Steel Structural Bolts, 150 ksi Minimum Tensile Strength 

A563, Specification for Carbon and Alloy Steel Nuts 

A572, Specification for High-Strength Low-Alloy Columbium-Vanadium Structural Steel 

A898, Straight Beam Ultrasonic Examination of Rolled Steel Structural Shapes 

A913, Specification for High-Strength Low-Alloy Steel Shapes of Structural Quality, Produced 

by Quenching and Self-Tempering Process 

A992, Standard Specification for Steel for Structural Shapes for Use in Building Framing 

E329, Standard Specification for Agencies Engaged in the Testing and/or Inspection of Material 

Used in Construction 

R-4
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

E543, Standard Practice for Agencies Performing Nondestructive Testing 

E548, Standard Guide for General Criteria Used for Evaluating Laboratory Competence 

E994, Standard Guide for Laboratory Accreditation Systems 

E1212, Standard Practice for Establishment and Maintenance of Quality Control Systems for 

Nondestructive Testing Agencies 

E1359, Standard Guide for Surveying Nondestructive Testing Agencies 

F436, Specification for Hardened Steel Washers 

F959, Specification for Compressible-Washer-Type Direct Tension Indicators for Use with 

Structural Fasteners 

F1554, Specification for Anchor Bolts, Steel, 36, 55, and 105 ksi Yield Strength 

F1852, Specification for “Twist-Off” Type Tension Control Structural Bolt/Nut/Washer 

Assemblies, Steel, Heat Treated, 120/105 ksi Minimum Tensile Strength 

AWS Reports, Specifications, and Codes. 

AWS reports are published by the American Welding Society, Miami, Florida, and are listed 

alphanumerically. 

AWS A2.4, Standard Symbols for Welding, Brazing, and Nondestructive Testing 

AWS A4.3, Standard Methods for Determination of the Diffusible Hydrogen Content of 

Martensitic, Bainitic, and Ferritic Steel Weld Metal Produced by Arc Welding 

ANSI/AWS A5.1-91, Specification for Carbon Steel Electrodes for Shielded Metal Arc Welding 

ANSI/AWS A5.18-93, Specification for Carbon Steel Electrodes and Rods for Gas Shielded Arc 

Welding 

ANSI/AWS A5.20-95, Specification for Carbon Steel Electrodes for Flux-Cored Arc Welding 

AWS, 1995, Presidential Task Group Report 

ANSI/AWS A5.5-96, Specification for Low-Alloy Steel Electrodes for Shielded Metal Arc 

Welding 

ANSI/AWS A5.28-96, Specification for Low-Alloy Steel Electrodes and Rods for Gas Shielded 

Arc Welding 

ANSI/AWS A5.23/A5.23M-97, Specification for Low-Alloy Steel Electrodes and Fluxes for 

Submerged Arc Welding 

ANSI/AWS A5.25/A5.25M-97, Specification for Carbon and Low-Alloy Steel Electrodes and 

Fluxes for Electroslag Welding 

ANSI/AWS A5.26/A5.26M-97, Specification for Carbon and Low-Alloy Steel Electrodes for 

Electrogas Welding 

ANSI/AWS A5.32/A5.32M-97, Specification for Welding Shielding Gases 

R-5
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

ANSI/AWS A5.17/A5.17M-97, Specification for Carbon Steel Electrodes and Fluxes for 

Submerged Arc Welding 

ANSI/AWS A5.29-98, Specification for Low-Alloy Steel Electrodes for Flux-Cored Arc Welding 

AWS D1.1-1998, 2000, Structural Welding Code – Steel 

AWS D1.3, Structural Welding Code 

AWS D1.4, Structural Welding Code 

AWS QC1, Standard for AWS Certification of Welding Inspectors 

FEMA Reports. 

FEMA reports are listed by report number. 

FEMA-178, 1992, NEHRP Handbook for the Seismic Evaluation of Existing Buildings, 

developed by the Building Seismic Safety Council for the Federal Emergency Management 

Agency, Washington, DC. 

FEMA-267, 1995, Interim Guidelines, Inspection, Evaluation, Repair, Upgrade and Design of 

Welded Moment Resisting Steel Structures, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Washington, DC. 

FEMA-267A, 1996, Interim Guidelines Advisory No. 1, prepared by the SAC Joint Venture for 

the Federal Emergency Management Agency, Washington, DC. 

FEMA-267B, 1999, Interim Guidelines Advisory No. 2, prepared by the SAC Joint Venture for 

the Federal Emergency Management Agency, Washington, DC. 

FEMA-273, 1997, NEHRP Guidelines for the Seismic Rehabilitation of Buildings, prepared by 

the Applied Technology Council for the Building Seismic Safety Council, published by the 

Federal Emergency Management Agency, Washington, DC. 

FEMA-274, 1997, NEHRP Commentary on the Guidelines for the Seismic Rehabilitation of 

Buildings, prepared by the Applied Technology Council for the Building Seismic Safety 

Council, published by the Federal Emergency Management Agency, Washington, DC. 

FEMA-302, 1997, NEHRP Recommended Provisions for Seismic Regulations for New Buildings 

and Other Structures, Part 1 – Provisions, prepared by the Building Seismic Safety Council 

for the Federal Emergency Management Agency, Washington, DC. 

FEMA-303, 1997, NEHRP Recommended Provisions for Seismic Regulations for New Buildings 

and Other Structures, Part 2 – Commentary, prepared by the Building Seismic Safety 

Council for the Federal Emergency Management Agency, Washington, DC. 

FEMA-310, 1998, Handbook for the Seismic Evaluation of Buildings – A Prestandard, prepared 

by the American Society of Civil Engineers for the Federal Emergency Management 

Agency, Washington, DC. 

FEMA-350, 2000, Recommended Seismic Design Criteria for New Steel Moment-Frame 

Buildings, prepared by the SAC Joint Venture for the Federal Emergency Management 

Agency, Washington, DC. 

R-6
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

FEMA-351, 2000, Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded 

Steel Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

FEMA-352, 2000, Recommended Postearthquake Evaluation and Repair Criteria for Welded 

Steel Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

FEMA-353, 2000, Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications, prepared by the SAC Joint Venture 

for the Federal Emergency Management Agency, Washington, DC. 

FEMA-354, 2000, A Policy Guide to Steel Moment-Frame Construction, prepared by the SAC 

Joint Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355A, 2000, State of the Art Report on Base Metals and Fracture, prepared by the SAC 

Joint Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355B, 2000, State of the Art Report on Welding and Inspection, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355C, 2000, State of the Art Report on Systems Performance of Steel Moment Frames 

Subject to Earthquake Ground Shaking, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

FEMA-355D, 2000, State of the Art Report on Connection Performance, prepared by the SAC 

Joint Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355E, 2000, State of the Art Report on Past Performance of Steel Moment-Frame 

Buildings in Earthquakes, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

FEMA-355F, 2000, State of the Art Report on Performance Prediction and Evaluation of Steel 

Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC Joint Venture Reports. 

SAC Joint Venture reports are listed by report number, except for SAC 2000a through 2000k; 

those entries that do not include a FEMA report number are published by the SAC Joint 

Venture. 

SAC 94-01, 1994, Proceedings of the Invitational Workshop on Steel Seismic Issues, Los 

Angeles, September 1994, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC 94-01, 1994b, Proceedings of the International Workshop on Steel Moment Frames, 

Sacramento, December, 1994, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC 95-01, 1995, Steel Moment Frame Connection Advisory No. 3, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Washington, DC. 

R-7
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

SAC 95-02, 1995, Interim Guidelines: Evaluation, Repair, Modification and Design of Welded 

Steel Moment Frame Structures, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Report No. FEMA-267, Washington, DC. 

SAC 95-03, 1995, Characterization of Ground Motions During the Northridge Earthquake of 

January 17, 1994, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC 95-04, 1995, Analytical and Field Investigations of Buildings Affected by the Northridge 

Earthquake of January 17, 1994, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

SAC 95-05, 1995, Parametric Analytic Investigations of Ground Motion and Structural 

Response, Northridge Earthquake of January 17, 1994, prepared by the SAC Joint Venture 

for the Federal Emergency Management Agency, Washington, DC. 

SAC 95-06, 1995, Technical Report: Surveys and Assessment of Damage to Buildings Affected 

by the Northridge Earthquake of January 17, 1994, prepared by the SAC Joint Venture for 

the Federal Emergency Management Agency, Washington, DC. 

SAC 95-07, 1995, Technical Report: Case Studies of Steel Moment-Frame Building 

Performance in the Northridge Earthquake of January 17, 1994, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Washington, DC. 

SAC 95-08, 1995, Experimental Investigations of Materials, Weldments and Nondestructive 

Examination Techniques, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC 95-09, 1995, Background Reports: Metallurgy, Fracture Mechanics, Welding, Moment 

Connections and Frame Systems Behavior, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Report No. FEMA-288, Washington, DC. 

SAC 96-01, 1996, Experimental Investigations of Beam-Column Subassemblages, Part 1 and 2, 

prepared by the SAC Joint Venture for the Federal Emergency Management Agency, 

Washington, DC. 

SAC 96-02, 1996, Connection Test Summaries, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Report No. FEMA-289, Washington, DC. 

SAC 96-03, 1997, Interim Guidelines Advisory No. 1 Supplement to FEMA-267 Interim 

Guidelines, prepared by the SAC Joint Venture for the Federal Emergency Management 

Agency, Report No. FEMA-267A, Washington, DC. 

SAC 98-PG, Update on the Seismic Safety of Steel Buildings – A Guide for Policy Makers, 

prepared by the SAC Joint Venture for the Federal Emergency Management Agency, 

Washington, DC. 

SAC 99-01, 1999, Interim Guidelines Advisory No. 2 Supplement to FEMA-267 Interim 

Guidelines, prepared by the SAC Joint Venture, for the Federal Emergency Management 

Agency, Report No. FEMA-267B, Washington, DC. 

R-8
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

SAC, 2000a, Recommended Seismic Design Criteria for New Steel Moment-Frame Buildings, 

prepared by the SAC Joint Venture for the Federal Emergency Management Agency, Report 

No. FEMA-350, Washington, D.C. 

SAC, 2000b, Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded Steel 

Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Report No. FEMA-351, Washington, D.C. 

SAC, 2000c, Recommended Postearthquake Evaluation and Repair Criteria for Welded Steel 

Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Report No. FEMA-352, Washington, D.C. 

SAC, 2000d, Recommended Specifications and Quality Assurance Guidelines for Steel MomentFrame Construction for Seismic Applications, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Report No. FEMA-353, Washington, D.C. 

SAC, 2000e, A Policy Guide to Steel Moment-Frame Construction, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-354, 

Washington, D.C. 

SAC, 2000f, State of the Art Report on Base Metals and Fracture, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355A, 

Washington, D.C. 

SAC, 2000g, State of the Art Report on Welding and Inspection, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355B, 

Washington, D.C. 

SAC, 2000h, State of the Art Report on Systems Performance, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355C, 

Washington, D.C. 

SAC, 2000i, State of the Art Report on Connection Performance, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355D, 

Washington, D.C. 

SAC, 2000j, State of the Art Report on Past Performance of Steel Moment-Frame Buildings in 

Earthquakes, prepared by the SAC Joint Venture for the Federal Emergency Management 

Agency, Report No. FEMA-355E, Washington, D.C. 

SAC, 2000k, State of the Art Report on Performance Prediction and Evaluation, prepared by the 

SAC Joint Venture for the Federal Emergency Management Agency, Report No. FEMA355F, Washington, D.C. 

SAC/BD-96/01, Selected Results from the SAC Phase 1 Beam-Column Connection Pre-Test 

Analyses, submissions from B. Maison, K. Kasai, and R. Dexter; and A. Ingraffea and G. 

Deierlein. 

SAC/BD-96/02, Summary Report on SAC Phase 1 - Task 7 Experimental Studies, by C. Roeder 

(a revised version of this document is published in Report No. SAC 96-01; the original is no 

longer available). 

SAC/BD-96/03, Selected Documents from the U.S.-Japan Workshop on Steel Fracture Issues. 

R-9
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

SAC/BD-96/04, Survey of Computer Programs for the Nonlinear Analysis of Steel Moment 

Frame Structures. 

SAC/BD-97/01, Through-Thickness Properties of Structural Steels, by J. Barsom and S. 

Korvink. 

SAC/BD-97/02, Protocol for Fabrication, Inspection, Testing, and Documentation of BeamColumn Connection Tests and Other Experimental Specimens, by P. Clark, K. Frank, H. 

Krawinkler, and R. Shaw. 

SAC/BD-97/03, Proposed Statistical and Reliability Framework for Comparing and Evaluating 

Predictive Models for Evaluation and Design, by Y.-K. Wen. 

SAC/BD-97/04, Development of Ground Motion Time Histories for Phase 2 of the FEMA/SAC 

Steel Project, by. P. Somerville, N. Smith, S. Punyamurthula, and J. Sun. 

SAC/BD-97/05, Finite Element Fracture Mechanics Investigation of Welded Beam-Column 

Connections, by W.-M. Chi, G. Deierlein, and A. Ingraffea. 

SAC/BD-98/01, Strength and Ductility of FR Welded-Bolted Connections, by S. El-Tawil, T. 

Mikesell, E. Vidarsson, and S. K. Kunnath. 

SAC/BD-98/02, Effects of Strain Hardening and Strain Aging on the K-Region of Structural 

Shapes, by J. Barsom and S. Korvink 

SAC/BD-98/03, Implementation Issues for Improved Seismic Design Criteria: Report on the 

Social, Economic, Policy and Political Issues Workshop by L.T. Tobin. 

SAC/BD-99/01, Parametric Study on the Effect of Ground Motion Intensity and Dynamic 

Characteristics on Seismic Demands in Steel Moment Resisting Frames by G. A. MacRae 

SAC/BD-99/01A, Appendix to: Parametric Study on the Effect of Ground Motion Intensity and 

Dynamic Characteristics on Seismic Demands in Steel Moment Resisting Frames by G. A. 

MacRae 

SAC/BD-99/02, Through-Thickness Strength and Ductility of Column Flange in Moment 

Connections by R. Dexter and M. Melendrez. 

SAC/BD-99/03, The Effects of Connection Fractures on Steel Moment Resisting Frame Seismic 

Demands and Safety by C. A. Cornell and N. Luco 

SAC/BD-99/04, Effects of Strength/Toughness Mismatch on Structural and Fracture Behaviors 

in Weldments by P. Dong, T. Kilinski, J. Zhang and F.W. Brust 

SAC/BD-99/05, Assessment of the Reliability of Available NDE Methods for Welded Joint and 

the Development of Improved UT Procedures by G. Gruber and G. Light 

SAC/BD-99/06, Prediction of Seismic Demands for SMRFs with Ductile Connections and 

Elements by A. Gupta and H. Krawinkler 

SAC/BD-99/07, Characterization of the Material Properties of Rolled Sections by T. K. Jaquess 

and K. Frank 

SAC/BD-99/08, Study of the Material Properties of the Web-Flange Intersection of Rolled 

Shapes by K. R. Miller and K. Frank 

R-10
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

SAC/BD-99/09, Investigation of Damage to WSMF Earthquakes other than Northridge by M. 

Phipps 

SAC/BD-99/10, Clarifying the Extent of Northridge Induced Weld Fracturing and Examining 

the Related Issue of UT Reliability by T. Paret 

SAC/BD-99/11, The Impact of Earthquakes on Welded steel Moment Frame Buildings: 

Experience in Past Earthquakes by P. Weinburg and J. Goltz 

SAC/BD-99/12, Assessment of the Benefits of Implementing the New Seismic Design Criteria 

and Inspection Procedures by H. A. Seligson and R. Eguchi 

SAC/BD-99/13, Earthquake Loss Estimation for WSMF Buildings, by C. A. Kircher 

SAC/BD-99/14, Simplified Loss Estimation for Pre-Northridge WSMF Buildings, by B. F. 

Maison and D. Bonowitz 

SAC/BD-99/15, Integrative Analytical Investigations on the Fracture Behavior of Welded 

Moment Resisting Connections, by G. G. Deierlein and W.-M. Chi 

SAC/BD-99/16, Seismic Performance of 3 and 9 Story Partially Restrained Moment Frame 

Buildings, by B. F. Maison and K. Kasai 

SAC/BD-99/17, Effects of Partially-Restrained Connection Stiffness and Strength on Frame 

Seismic Performance, by K. Kasai, B. F. Maison, and A. Mayangarum 

SAC/BD-99/18, Effects of Hysteretic Deterioration Characteristics on Seismic Response of 

Moment Resisting Steel Structures, by F. Naeim, K. Skliros, A. M. Reinhorn and M.V. 

Sivaselvan 

SAC/BD-99/19, Cyclic Instability of Steel Moment Connections with Reduced Beam Section, by 

C.-M. Uang and C.-C. Fan 

SAC/BD-99/20, Local and Lateral-Torsion Buckling of Wide Flange Beams, by L. 

Kwasniewski, B. Stojadinovic, and S. C. Goel 

SAC/BD-99/21, Elastic Models for Predicting Building Performance, by X. Duan and J. C. 

Anderson 

SAC/BD-99/22, Reliability-Based Seismic Performance Evaluation of Steel Frame Buildings 

Using Nonlinear Static Analysis Methods, by G. C. Hart and M. J. Skokan 

SAC/BD-99/23, Failure Analysis of Welded Beam to Column Connections, by J. M. Barsom 

SAC/BD-99/24, Weld Acceptance Criteria for Seismically-Loaded Welded Connections, by W. 

Mohr 

SAC/BD-00/01, Parametric Tests on Unreinforced Connections, by K.-H. Lee, B. Stojadinovic, 

S. C. Goel, A. G. Margarian, J. Choi, A. Wongkaew, B. P. Reyher, and D.-Y, Lee 

SAC/BD-00/02, Parametric Tests on the Free Flange Connections, by J. Choi, B. Stojadinovic, 

and S. C. Goel 

SAC/BD-00/03, Cyclic Tests on Simple Connections Including Effects of the Slab, by J. Liu and 

A. Astaneh-Asl 

R-11
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

SAC/BD-00/04, Tests on Bolted Connections, by J. Swanson, R. Leon and J. Smallridge 

SAC/BD-00/05, Bolted Flange Plate Connections, by S. P. Schneider and I. Teeraparbwong 

SAC/BD-00/06, Round Robin Testing of Ultrasonic Testing Technicians, by R. E. Shaw, Jr. 

SAC/BD-00/07, Dynamic Tension Tests of Simulated Welded Beam Flange Connections, by J. 

M. Ricles, C. Mao, E. J. Kaufmann, L.-W. Lu, and J. W. Fisher 

SAC/BD-00/08, Design of Steel Moment Frame Model Buildings in Los Angeles, Seattle and 

Boston, by P. Clark 

SAC/BD-00/09, Benchmarking of Analysis Programs for SMRF System Performance Studies, by 

A. G. and H. Krawinkler 

SAC/BD-00/10, Loading Histories for Seismic Performance Testing of SMRF Components and 

Assemblies, by H. Krawinkler, A. Gupta, R. Medina and N. Luco 

SAC/BD-00/11, Development of Improved Post-Earthquake Inspection Procedures for Steel 

Moment Frame Buildings, by P. Clark 

SAC/BD-00/12, Evaluation of the Effect of Welding Procedure on the Mechanical Properties of 

FCAW-S and SMAW Weld Metal Used in the Construction of Seismic Moment Frames, by 

M. Q. Johnson 

SAC/BD-00/13, Preliminary Evaluation of Heat Affected Zone Toughness in Structural Shapes 

Used in the Construction of Seismic Moment Frames, by M. Q. Johnson 

SAC/BD-00/14, Evaluation of Mechanical Properties in Full-Scale Connections and 

Recommended Minimum Weld Toughness for Moment Resisting Frames, by M. Q. Johnson, 

W. Mohr, and J. Barsom 

SAC/BD-00/15, Simplified Design Models for Predicting the Seismic Performance of Steel 

Moment Frame Connections, by C. Roeder, R.G. Coons, and M. Hoit 

SAC/BD-00/16, SAC Phase 2 Test Plan, by C. Roeder 

SAC/ BD-00/17, Behavior and Design of Radius-Cut, Reduced Beam Section Connections, by 

M. Engelhardt, G. Fry, S. Johns, M. Venti, and S. Holliday 

SAC/BD-00/18, Test of a Free Flange Connection with a Composite Floor Slab, by M. Venti 

and M. Engelhardt 

SAC/BD-00/19, Cyclic Testing of a Free Flange Moment Connection by C. Gilton, B. Chi, and 

C. M. Uang 

SAC/BD-00/20, Improvement of Welded Connections Using Fracture Tough Overlays, by James 

Anderson, J. Duan, P. Maranian, and Y. Xiao 

SAC/BD-00/21, Cyclic Testing of Bolted Moment End-Plate Connections, by T. Murray and E. 

Sumner 

SAC/BD-00/22, Cyclic Response of RBS Moment Connections: Loading Sequence and Lateral 

Bracing Effects, by Q.S. Yu, C. Gilton, and C. M. Uang 

R-12
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

SAC/BD-00/23, Cyclic Response of RBS Moment Connections: Weak Axis Configuration and 

Deep Column Effects, by C. Gilton, B. Chi, and C. M. Uang 

SAC/BD-00/24, Development and Evaluation of Improved Details for Ductile Welded 

Unreinforced Flange Connections, by J.M. Ricles, C. Mao, L.W. Lu, and J. Fisher 

SAC/BD-00/25, Performance Prediction and Evaluation of Steel Special Moment Frames for 

Seismic Loads, by K. Lee and D. A. Foutch 

SAC/BD-00/26, Performance Prediction and Evaluation of Low Ductility Steel Moment Frames 

for Seismic Loads, by S. Yun and D. A. Foutch 

SAC/BD-00/27, Steel Moment Resisting Connections Reinforced with Cover and Flange Plates, 

by T. Kim, A.S. Whittaker, V.V. Bertero, and A.S.J. Gilani 

SAC/BD-00/28, Failure of a Column K-Area Fracture, by J.M. Barsom and J.V. Pellegrino 

SAC/BD-00/29, Inspection Technology Workshop, by R. E. Shaw, Jr. 

Acronyms. 

A, acceleration response

ACAG, air carbon arc gouging

ACIL, American Council of Independent

Laboratories 

AISC, American Institute for Steel 

Construction 

ANSI, American National Standards Institute 

API, American Petroleum Institute 

ASNT, American Society for Nondestructive 

Testing 

ASTM, American Society for Testing and 

Materials 

ATC, Applied Technology Council 

A2LA, American Association for Laboratory 

Accreditation 

AWS, American Welding Society 

BB, Bolted Bracket (connection) 

BFP, Bolted Flange Plates (connection) 

BOCA, Building Officials and Code 

Administrators 

BSEP, Bolted Stiffened End Plate 

(connection) 

BUEP, Bolted Unstiffened End Plate 

(connection) 

CAC-A, air carbon arc cutting 

CAWI, Certified Associate Welding Inspector 

CJP, complete joint penetration (weld) 

CP, Collapse Prevention (performance level) 

CUREe, California Universities for Research 

in Earthquake Engineering 

CVN, Charpy V-notch 

CWI, Certified Welding Inspector 

D, displacement response 

DST, Double Split Tee (connection) 

DTI, Direct Tension Indicator 

EGW, electrogas welding 

ELF, equivalent lateral force 

ESW, electroslag welding 

FCAW-S, flux-cored arc welding – selfshielded 

FCAW-G, flux-cored arc welding – gasshielded 

FEMA, Federal Emergency Management 

Agency 

FF, Free Flange (connection) 

FR, fully restrained (connection) 

GMAW, gas metal arc welding 

GTAW, gas tungsten arc welding 

HAZ, heat-affected zone 

IBC, International Building Code 

ICBO, International Conference of Building 

Officials 

ICC, International Code Council 

IMF, Intermediate Moment Frame 

IO, Immediate Occupancy (performance 

level) 

R-13
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

ISO, International Standardization 

Organization 

IWURF, Improved Welded Unreinforced 

Flange (connection) 

L, longitudinal 

LDP, Linear Dynamic Procedure 

LRFD, load and resistance-factor design 

LS, Life Safety (performance level) 

LSP, Linear Static Procedure 

MCE, Maximum Considered Earthquake 

MMI, Modified Mercalli Intensity 

MRS, modal response spectrum 

MRSF, steel moment frame 

MT, magnetic particle testing 

NBC, National Building Code 

NDE, nondestructive examination 

NDP, Nonlinear Dynamic Procedure 

NDT, nondestructive testing 

NEHRP, National Earthquake Hazard 

Reduction Program 

NES, National Evaluation Services 

NSP, Nonlinear Static Procedure 

NVLAP, National Volunteer Laboratory 

Accreditation Program 

OMF, Ordinary Moment Frame 

PGA, peak ground acceleration 

PJP, partial joint penetration (weld) 

PIDR, pseudo interstory drift ratio 

PQR, Performance Qualification Record 

PR, partially restrained (connection) 

PT, liquid dye penetrant testing 

PWHT, postweld heat treatment 

PZ, panel zone 

QA, quality assurance 

QC, quality control 

QCP, Quality Control Plan, Quality 

Certification Program 

RBS, Reduced Beam Section (connection) 

RCSC, Research Council for Structural 

Connections 

RT, radiographic testing 

SAC, the SAC Joint Venture; a partnership of 

the Structural Engineers Association of 

California, the Applied Technology 

Council, and California Universities for 

Research in Earthquake Engineering 

SAW, submerged arc welding

SBC, Standard Building Code

SBCCI, Southern Building Code Congress

International 

SCWI, Senior Certified Welding Inspector 

SEAOC, Structural Engineers Association of 

California 

SFRS, seismic-force-resisting system 

SMAW, shielded metal arc welding 

SMF, Special Moment Frame 

SP, Side Plate (connection) 

SUG, Seismic Use Group 

SW, Slotted Web (connection) 

T, transverse 

TIGW, tungsten inert gas welding 

UBC, Uniform Building Code 

UT, ultrasonic testing 

VI, visual inspection 

WBH, Welded Bottom Haunch (connection) 

WCPF, Welded Cover Plate Flange 

(connection) 

WFP, Welded Flange Plate (connection) 

WPQR, Welding Performance Qualification 

Record 

WPS, Welding Procedure Specification 

WSMF, welded steel moment frame 

WT, Welded Top Haunch (connection) 

WTBH, Welded Top and Bottom Haunch 

(connection) 

WUF-B, Welded Unreinforced Flanges – 

Bolted Web (connection) 

WUF-W, Welded Unreinforced Flanges – 

Welded Web (connection) 

R-14
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings SAC Project Participants 

SAC Phase II Project Participants 

FEMA Project Officer FEMA Technical Advisor 

Michael Mahoney Robert D. Hanson

Federal Emergency Management Agency Federal Emergency Management Agency

500 C St. SW, Room 404 DFO Room 353

Washington, DC 20472 P.O. Box 6020

Pasadena, CA 91102-6020 

Joint Venture Management Committee (JVMC) 

William T. Holmes, Chair

Rutherford and Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

Edwin T. Huston

Smith & Huston, Inc.

8618 Roosevelt Way NE

Seattle, WA 98115

Robert Reitherman

California Universities for Research in

Earthquake Engineering 

1301 South 46th St. 

Richmond, CA 94804 

Christopher Rojahn

Applied Technology Council

555 Twin Dolphin Dr., Suite 550

Redwood City, CA 94065

Arthur E. Ross

Cole/Yee/Shubert & Associates

2500 Venture Oaks Way, Suite 100

Sacramento, CA 95833

Robin Shepherd

Earthquake Damage Analysis Corporation

40585 Lakeview Drive, Suite 1B

P.O. Box 1967

Big Bear Lake, CA 92315

Project Management Committee (PMC) 

Stephen A. Mahin, Project Manager

Pacific Earthquake Engr. Research Center

University of California

Berkeley, CA 94720

Ronald O. Hamburger, Project Director for

Project Development

EQE International

1111 Broadway, 10th Floor

Oakland, CA 94607-5500

James O. Malley, Project Director for

Topical Investigations

Degenkolb Engineers

225 Bush St., Suite 1000

San Francisco, CA 94104-1737

William T. Holmes, JVMC

Rutherford and Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

Christopher Rojahn, JVMC

Applied Technology Council

555 Twin Dolphin Dr., Suite 550

Redwood City, CA 94065

Robin Shepherd, JVMC

Earthquake Damage Analysis Corporation

40585 Lakeview Drive, Suite 1B

P.O. Box 1967

Big Bear Lake, CA 92315

S-1
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

SAC Project Participants Steel Moment-Frame Buildings 

Peter W. Clark, Technical Assistant to PMC

SAC Steel Project Technical Office

1301 South 46th St.

Richmond, CA 94804

Project Administration 

Allen Paul Goldstein, Project Administrator Lori Campbell, Assistant to the Project 

A.P. Goldstein Associates

1621B 13th Street

Sacramento, CA 95814

Lee Adler

Structural Engineers Association of

California 

1730 I Street, Ste. 240 

Sacramento, CA 95814 

Administrator 

1621 B 13th Street 

Sacramento, CA 95628 

Project Oversight Committee (POC) 

William J. Hall, Chair

3105 Valley Brook Dr.

Champaign, IL 61821

Shirin Ader

International Conference of Building

Officials 

5360 Workman Mill Rd. 

Whittier, CA 90601-2298 

John M. Barsom

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

Roger Ferch

Herrick Corporation

7021 Koll Center Parkway

P.O Box 9125

Pleasanton, CA 94566-9125

Theodore V. Galambos

University of Minnesota

122 CE Building, 500 Pillsbury Dr. SE

Minnneapolis, MN 55455

John L. Gross

National Institute of Stds. & Technology

Building and Fire Research Lab,

Building 226, Room B158

Gaithersburg, MD 20899

James R. Harris

J.R. Harris and Co.

1580 Lincoln St., Suite 550

Denver, CO 80203-1509

Richard Holguin

520 Kathryn Ct.

Nipomo, CA 93444

Nestor Iwankiw

American Institute of Steel Construction

One East Wacker Dr., Suite 3100

Chicago, IL 60601-2001

Roy Johnston

Brandow & Johnston Associates

1600 West 3rd St.

Los Angeles, CA 90017

S-2
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings SAC Project Participants 

Leonard Joseph

Thornton-Tomassetti Engineers

641 6th Ave., 7th Floor

New York, NY 10011

Duane K. Miller

The Lincoln Electric Company

22801 St. Clair Ave.

Cleveland, OH 44117-1194

John Theiss

EQE/Theiss Engineers

1848 Lackland Hills Parkway

St. Louis, MO 63146-3572

John H. Wiggins

J.H. Wiggins Company

1650 South Pacific Coast Hwy, Suite 311

Redondo Beach, CA 90277

Team Leaders for Topical Investigations 

Douglas A. Foutch

University of Illinois

MC-250, 205 N. Mathews Ave.

3129 Newmark Civil Engineering Lab

Urbana, IL 61801

Karl H. Frank

University of Texas at Austin

10100 Bornet Rd.

Ferguson Lab, P.R.C. #177

Austin, TX 78758

Matthew Johnson

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

Helmut Krawinkler

Department of Civil Engineering

Stanford University

Stanford, CA 94305

Charles W. Roeder

University of Washington

233-B More Hall FX-10

Dept. of Building and Safety

Seattle, WA 98195-2700

L. Thomas Tobin

Tobin and Associates

134 California Ave.

Mill Valley, CA 94941

Lead Guideline Writers 

John D. Hooper

Skilling Ward Magnusson Barkshire, Inc.

1301 Fifth Avenue, Suite 3200

Seattle, WA 98101-2699

Lawrence D. Reaveley

University of Utah

Civil Engineering Dept.

3220 Merrill Engineering Building

Salt Lake City, UT 84112

Thomas A. Sabol

Englekirk & Sabol Consulting Engineers

P.O. Box 77-D

Los Angeles, CA 90007

C. Mark Saunders

Rutherford & Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

Robert E. Shaw

Steel Structures Technology Center, Inc.

42400 W Nine Mile Road

Novi, MI 48375-4132

Raymond H. R. Tide

Wiss, Janney, Elstner Associates, Inc.

330 Pfingsten Road

Northbrook, IL 60062-2095

S-3
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

SAC Project Participants Steel Moment-Frame Buildings 

C. Allin Cornell, Associate Guideline Writer

Stanford University

Terman Engineering Center

Stanford, CA 94305-4020

Technical Advisory Panel (TAP) for Materials and Fracture 

John M. Barsom, POC

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

Serge Bouchard*

TradeARBED

825 Third Avenue, 35th Floor

New York, NY 10022

Michael F. Engestrom*

Nucor-Yamato Steel

P.O. Box 678

Frederick, MD 21705-0678

Karl H. Frank, Team Leader

University of Texas at Austin

10100 Bornet Rd.

Ferguson Lab, P.R.C. #177

Austin, TX 78758

Nestor Iwankiw*

American Institute of Steel Construction

One East Wacker Dr., Suite 3100

Chicago, IL 60601-2001

Dean C. Krouse*

705 Pine Top Drive

Bethelem, PA 18017

Frederick V. Lawrence

University of Illinois at Urbana-Champaign

205 N. Mathews Ave.

Room 2129 Newmark Lab

Urbana, IL 61801

Robert F. Preece

Preece, Goudie & Associates

100 Bush St., Suite 410

San Francisco, CA 94104

Raymond H. R. Tide, Guideline Writer

Wiss, Janney, Elstner Associates, Inc.

330 Pfingsten Road

Northbrook, IL 60062-2095

TAP for Welding and Inspection 

John M. Barsom

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

John W. Fisher

Lehigh University

117 ATLSS Drive

Bethlehem, PA 18015-4729

J. Ernesto Indacochea

University of Illinois at Chicago

Civil and Materials Engineering (mc 246)

842 West Taylor Street

Chicago, IL 60607

Matthew Johnson, Team Leader

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

S-4
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings SAC Project Participants 

David Long

PDM Strocal, Inc.

2324 Navy Drive

Stockton, CA 95206

Duane K. Miller, POC

The Lincoln Electric Company

22801 St. Clair Ave.

Cleveland, OH 44117-1194

Robert Pyle*

AISC Marketing

10101 South State Street

Sandy, Utah 84070

Douglas Rees-Evans*

Steel Dynamics, Inc.

Structural Mill Division

2601 County Road 700 East

Columbia City, IN 46725

Richard I. Seals

P.O. Box 11327

Berkeley, CA 94712-2327

Robert E. Shaw, Guideline Writer

Steel Structures Technology Center, Inc.

42400 W Nine Mile Road

Novi, MI 48375-4132

TAP for Connection Performance 

Charlie Carter*

American Institute of Steel Construction

One East Wacker Drive, Suite 3100

Chicago, IL 60601-2001

Robert H. Dodds

University of Illinois at Urbana-Champaign

205 N. Mathews Ave.

2129 Newmark Lab

Urbana, IL 61801

Roger Ferch, POC

Herrick Corporation

7021 Koll Center Parkway

P.O Box 9125

Pleasanton, CA 94566-9125

John D. Hooper, Guideline Writer

Skilling Ward Magnusson Barkshire, Inc.

1301 Fifth Avenue, Suite 3200

Seattle, WA 98101-2699

Egor Popov

University of California at Berkeley

Department of Civil and Environmental

Engineering, Davis Hall 

Berkeley, CA 94720 

Steve Powell*

SME Steel Contractors

5955 W. Wells Park Rd.

West Jordan, UT 84088

Charles W. Roeder, Team Leader

University of Washington

233-B More Hall FX-10

Dept. of Building and Safety

Seattle, WA 98195-2700

Stanley T. Rolfe

University of Kansas

Civil Engineering Department

2006 Learned Hall

Lawrence, KS 66045-2225

Rick Wilkinson*

Gayle Manufacturing Company

1455 East Kentucky

Woodland, CA 95695

S-5
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

SAC Project Participants Steel Moment-Frame Buildings 

TAP for System Performance 

Jacques Cattan*

American Institute of Steel Construction

One East Wacker Drive, Suite 3100

Chicago, IL 60601-2001

Gary C. Hart

Hart Consultant Group

The Water Garden, Ste. 670E

2425 Olympic Blvd.

Santa Monica, CA 90404-4030

Y. Henry Huang*

Los Angeles County Dept. of Public Works

900 S. Fremont Avenue, 8th Floor

Alhambra, CA 91803

Helmut Krawinkler, Team Leader

Department of Civil Engineering

Stanford University

Stanford, CA 94305

Dennis Randall*

SME Steel Contractors

5955 West Wells Park Road

West Jordan, UT 84088

Andrei M. Reinhorn

State University of New York at Buffalo

Civil Engineering Department

231 Ketter Hall

Buffalo, NY 14260

Arthur E. Ross, JVMC

Cole/Yee/Shubert & Associates

2500 Venture Oaks Way, Suite 100

Sacramento, CA 95833

C. Mark Saunders, Guideline Writer

Rutherford & Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

W. Lee Shoemaker*

Metal Building Manufacturers Association

1300 Summer Avenue

Cleveland, OH 44115

John Theiss, POC

EQE/Theiss Engineers

1848 Lackland Hills Parkway

St. Louis, MO 63146-3572

TAP for Performance Prediction and Evaluation 

Vitelmo V. Bertero

University of California at Berkeley

Pacific Earthquake Engr. Research Center

1301 S. 46th St.

Richmond, CA 94804

Bruce R. Ellingwood

Johns Hopkins University

Department of Civil Engineering

3400 N. Charles St.

Baltimore, MD 21218

Douglas A. Foutch, Team Leader

University of Illinois

MC-250, 205 N. Mathews Ave.

3129 Newmark Civil Engineering Lab

Urbana, IL 61801

Theodore V. Galambos, POC

University of Minnesota

122 CE Building, 500 Pillsbury Dr. SE

Minnneapolis, MN 55455

Lawrence G. Griffis

Walter P. Moore & Associates

3131 Eastside, Second Floor

Houston, TX 77098

Edwin T. Huston, JVMC

Smith & Huston, Inc.

8618 Roosevelt Way NE

Seattle, WA 98115

S-6
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings SAC Project Participants 

Harry Martin*

American Iron and Steel Institute

11899 Edgewood Road, Suite G

Auburn, CA 95603

Thomas A. Sabol, Guideline Writer

Englekirk & Sabol Consulting Engineers

Tom Schlafly*

American Institute of Steel Construction

One East Wacker Drive, Suite 3100

Chicago, IL 60601-2001

P.O. Box 77-D

Los Angeles, CA 90007

NormAbrahamson

Pacific Gas & Electric

P.O. Box 770000, MC N4C

San Francisco, CA 94177

C.B. Crouse

URS – Dames and Moore

2025 First Avenue, Suite 500

Seattle, WA 98121

Technical Advisors 

Robert Kennedy

RPK Structural Mechanics Consultants

18971 Villa Terr

Yorba Linda, CA 92886

Social Economic and Policy Panel 

Martha Cox-Nitikman

Building and Owners and Managers

Association, Los Angeles 

700 South Flower, Suite 2325 

Los Angeles, CA 90017 

Karl Deppe

27502 Fawnskin Dr.

Rancho Palos Verdes, CA 90275

Eugene Lecomte

Institute for Business and Home Safety

6 Sheffield Drive

Billerica, MA 01821

James Madison

Attorney at Law, Mediator and Arbitrator

750 Menlo Avenue, Suite 250

Menlo Park, CA 94025

Alan Merson

Morley Builders

2901 28th Street, Suite 100

Santa Monica, CA 90405

Joanne Nigg

University of Delaware

Disaster Research Center

Newark, DE 19716

William Petak

University of Southern California

Lewis Hall, Room 201

650 Childs Way

Los Angeles, CA 90089

Francine Rabinovitz

Hamilton, Rabinovitz and Alschuler

1990 South Bundy Drive, Suite 777

Los Angeles, CA 90025

S-7
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

SAC Project Participants Steel Moment-Frame Buildings 

Dennis Randall

SME Steel Contractors

5955 West Wells Park Road

West Jordan, UT 84088

David Ratterman

Stites and Harbison

400 West Market St., Suite 1800

Louisville, KY 40202-3352

L. Thomas Tobin, Panel Coordinator

134 California Ave.

Mill Valley, CA 94941

Stephen Toth

TIAA-CREF

730 Third Avenue

New York, NY 10017-3206

John H. Wiggins, POC

J.H. Wiggins Company

1650 South Pacific Coast Hwy, Suite 311

Redondo Beach, CA 90277

Performance of Steel Buildings in Past Earthquakes Subcontractors 

David Bonowitz

887 Bush, No. 610

San Francisco, CA 94108

Peter Clark

SAC Steel Project Technical Office

1301 South 46th St.

Richmond, CA 94804

Michael Durkin

Michael Durkin & Associates

22955 Leanora Dr.

Woodland Hills, CA 91367

James Goltz

California Institute of Technology

Office of Earthquake Programs

Mail Code 252-21

Pasadena, CA 91125

Bruce Maison

7309 Lynn Ave

Elcerrito, CA 94530

Peter Maranian

Brandow & Johnston Associates

1660 West Third Street

Los Angeles, CA 90017

Terrence Paret

Wiss Janney Elstner Associates, Inc.

2200 Powell St. Suite 925

Emeryville, CA 94602

Maryann Phipps

Degenkolb Engineers

225 Bush Street, Suite 1000

San Francisco, CA 94104

Allan Porush

Dames & Moore

911 Wilshire Blvd., Suite 700

Los Angeles, CA 90017

Access Current Knowledge Subcontractors 

David Bonowitz Stephen Liu

887 Bush , No. 610 Colorado School of Mines

San Francisco, CA 94108 Mathematics and Computer Science

Department 

Golden, CO 80401 

S-8
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings SAC Project Participants 

Materials and Fracture Subcontractors 

Robert Dexter Karl H. Frank

University of Minnesota University of Texas at Austin

122 Civil Engineering Building 10100 Bornet Rd.

500 Pillsbury Drive SE Ferguson Lab, P.R.C. #177

Minneapolis, MN 55455-0116 Austin, TX 78758

Welding and Inspection Subcontractors 

Pingsha Dong / Tom Kilinski Glenn M. Light / George Gruber 

Center for Welded Structures Research Southwest Research Institute 

Battelle Memorial Institute

501 King Avenue

Columbus, OH 43201-2693

Matthew Johnson

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

6220 Culebra Road, P. O. Drawer 28510

San Antonio, TX 78228-0510

William C. Mohr

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

Connection Performance Subcontractors 

Gregory Deierlein

Stanford University

Terman Engineering Center

Department of Civil and Enviromental Engr.

Stanford, CA 94305-4020

Charles W. Roeder

University of Washington

233-B More Hall FX-10

Seattle, WA 98195-2700

Sherif El-Tawil / Sashi Kunnath

University of Central Florida

Civil and Environmental Engr. Department

Orlando, FL. 32816-2450

Anthony Ingraffea

Cornell University

School of Civil Engineering

363 Hollister Hall

Ithaca, NY 14853

System Performance Subcontractors 

Paul Somerville

Woodward-Clyde Federal Services

566 El Dorado St., Suite 100

Pasadena, CA 91101-2560

Farzad Naeim

John A. Martin & Associates

1212 S. Flower Ave.

Los Angeles, CA 90015

Andrei M. Reinhorn

State University of New York at Buffalo

Civil Engineering Department

231 Ketter Hall

Buffalo, NY 14260

C. Allin Cornell

Stanford University

Terman Engineering Center

Stanford, CA 94305-4020

S-9
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

SAC Project Participants Steel Moment-Frame Buildings 

Helmut Krawinkler

Dept. of Civil Engineering

Stanford University

Stanford, CA 94305

Gregory MacRae

University of Washington

Civil Engineering Department

Seattle, WA 98195-2700

Kazuhiko Kasai

Tokyo Institute of Technology

Structural Engineering Research Center

Nagatsuta, Midori-Ku

Yokohama 226-8503, JAPAN

Bruce F. Maison

7309 Lynn Avenue

El Cerrito, CA 94530

Performance Prediction and Evaluation Subcontractors 

James Anderson

University of Southern California

Civil Engineering Department

Los Angeles, CA 90089-2531

Douglas A. Foutch

University of Illinois

MC-250, 205 N. Mathews Ave.

3129 Newmark Civil Engineering Lab

Urbana, IL 61801

Gary C. Hart

Department of Civil and Environmental

Engineering 

University of California 

Los Angeles, CA 90095 

Y.K. Wen

University of Illinois

3129 Newmark Civil Engineering Lab

205 N. Mathews Ave.

Urbana, IL 61801

Testing Subcontractors 

Subhash Goel / Bozidar Stojadinovic

University of Michigan

Civil Engineering Department

Ann Arbor, MI 48109

Roberto Leon

Georgia Institute of Technology

School of Civil & Environmental Engr.

790 Atlantic Ave.

Atlanta, GA 30332-0355

Vitelmo V. Bertero / Andrew Whittaker

UC Berkeley

Pacific Earthquake Engr. Research Center

1301 S. 46th St.

Richmond, CA 94804

Thomas Murray

Virginia Tech, Dept. of Civil Engineering

200 Patton Hall

Blacksburg, VA 24061

James M. Ricles / Le-Wu Lu

Lehigh University

c/o ATLSS Center

117 ATLSS Drive, H Building

Bethlehem, PA 18015-4729

John M. Barsom

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

S-10
Recommended Postearthquake Evaluation 

And Repair Criteria for Welded FEMA-352 

Steel Moment-Frame Buildings SAC Project Participants 

Hassan Astaneh

University of California at Berkeley

Dept. of Civil and Environmental Engr.

781 Davis Hall

Berkeley, CA 94720

Michael Engelhardt

University of Texas at Austin

Ferguson Laboratory

10100 Burnet Road, Building 177

Austin, TX 78712-1076

Gary T. Fry

Texas A&M University

Department of Civil Engineering

Constructed Facilities Division, CE/TTI

Building, Room 710D 

College Station, TX 77843-3136 

Chia-Ming Uang

University of California at San Diego

Dept. of AMES, Division of Structural Engr.

409 University Center

La Jolla, California 92093-0085

Stephen Schneider

University of Ilinois at Urbana-Champaign

3106 Newmark Civil Engr. Lab, MC-250

205 N. Mathews Avenue

Urbana, IL 61801

Matthew Johnson

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

James Anderson

University of Southern California

Civil Engineering Department

Los Angeles, CA 90089-2531

Bozidar Stojadinovic

Dept. of Civil & Environmental Engr.

University of California

Berkeley, CA 94720

Inspection Procedure Consultants 

Thomas Albert

Digiray Corporation

2235 Omega Road, No. 3

San Ramon, CA 94583

Randal Fong

Automated Inspection Systems, Inc.

4861 Sunrise Drive, Suite 101

Martinez, CA 94553

Andre Lamarre

R.D Tech, Inc.

1200 St. Jean Baptiste, Suite 120

Quebec City, Quebec, Canada G2ZE 5E8

Glenn Light

Southwest Research Institute

6220 Culebra Road

San Antonio, TX 78228

Andrey Mishin

AS & E High Energy Systems

330 Keller Street, Building 101

Santa Clara, CA 95054

Robert Shaw

Steel Structures Technology Center, Inc.

42400 W. Nine Mile Road

Novi, MI 48375-4132

Carlos Ventura

Dept of Civil Engineering

University of British Columbia

2324 Main Hall

Vancouver, BC, Canada V6T 1Z4

S-11
Recommended Postearthquake Evaluation 

FEMA-352 And Repair Criteria for Welded 

SAC Project Participants Steel Moment-Frame Buildings 

Guideline Trial Applications Subcontractors 

John Hopper Lawrence Novak

Skilling Ward Magnusson Barkshire, Inc. Skidmore, Owings, and Merrill

1301 Fifth Avenue, Suite 320 

Seattle WA 98101-2699 

Leonard Joseph 

Thornton-Tomassetti Engineers 

641 6th Avenue, 7th Floor 

New York, NY 10011 

224 S. Michigan Ave, Suite 1000 

Chicago, IL 60604 

Maryann Phipps 

Degenkolb Engineers 

225 Bush Street, Suite 1000 

San Francisco, CA 94104 

Economic and Social Impact Study Subcontractors 

Ronald Eguchi

EQE Engineering and Design

300 Commerce Dr., Ste. 200

Irvine, CA 92602

Martin Gordon / Peter Morris

Adamson Associates

170 Columbus Avenue

San Francisco, CA 94133

Richard Henige

Lemessurier Consultants Inc.

675 Massachusetts Ave.

Cambridge, MA 02139-3309

Charles Kircher

Charles Kircher & Associates

1121 San Antonio Road, Suite D-202

Palo Alto, CA 94303

Lizandro Mercado

Brandow & Johnston Associates

1600 West 3rd St.

Los Angeles, CA 90017

Greg Schindler

KPFF Consulting Engineers

1201 3rd Ave.

Seattle, WA 98101-3000

Report Production and Administrative Services 

A. Gerald Brady, Technical Editor Carol Cameron, Publications Coordinator

Patricia A. Mork, Administrative Asst. Ericka Holmon, Admin. Assistant

Peter N. Mork, Computer Specialist California Universities for Research in

Bernadette A. Mosby, Operations Admin. Earthquake Engineering

Michelle S. Schwartzbach, Pub. Specialist 1301 S. 46th Street

Applied Technology Council Richmond, CA 94804

555 Twin Dolphin Drive, Suite 550

Redwood City, CA 94065

*indicates industrial or organizational contact representative 

S-12