DISCLAIMER 

This document provides recommended criteria for the design of steel moment-frame buildings to 

resist the effects of earthquakes. These recommendations were developed by practicing engineers, 

based on professional judgment and experience, and by a program of laboratory, field and analytical 

research. While every effort has been made to solicit comments from a broad selection of the 

affected parties, this is not a consensus document. It is primarily intended as a resource document 

for organizations with appropriate consensus processes for the development of future design 

standards and building code provisions. No warranty is offered, with regard to the 

recommendations contained herein, either by the Federal Emergency Management Agency, 

the SAC Joint Venture, the individual Joint Venture partners, or their directors, members or 

employees. These organizations and their employees do not assume any legal liability or 

responsibility for the accuracy, completeness, or usefulness of any of the information, 

products or processes included in this publication. The reader is cautioned to review 

carefully the material presented herein and exercise independent judgment as to its 

suitability for application to specific engineering projects. These recommended criteria have 

been prepared by the SAC Joint Venture with funding provided by the Federal Emergency 

Management Agency, under contract number EMW-95-C-4770. 

Cover Art. The beam-column connection assembly shown on the cover depicts the standard 

detailing used in welded steel moment-frame construction prior to the 1994 Northridge 

earthquake. This connection detail was routinely specified by designers in the period 1970-1994 

and was prescribed by the Uniform Building Code for seismic applications during the period 

1985-1994. It is no longer considered to be an acceptable design for seismic applications. 

Following the Northridge earthquake, it was discovered that many of these beam-column 

connections had experienced brittle fractures at the joints between the beam flanges and column 

flanges. 
Recommended Seismic Design Criteria for New Steel 

Moment-Frame Buildings 

SAC Joint Venture 

A partnership of

Structural Engineers Association of California (SEAOC)

Applied Technology Council (ATC)

California Universities for Research in Earthquake Engineering (CUREe)

Prepared for SAC Joint Venture Partnership by

Guidelines Development Committee

Ronald O. Hamburger, Chair 

John D. Hooper Thomas Sabol 

Robert Shaw C. Mark Saunders 

Lawrence D. Reaveley Raymond H. R. Tide 

Project Oversight Committee 

William J. Hall, Chair 

Shirin Ader

John M. Barsom

Roger Ferch

Theodore V. Galambos

John Gross 

James R. Harris 

Richard Holguin 

Nestor Iwankiw 

Roy G. Johnston 

Leonard Joseph 

Duane K. Miller 

John Theiss 

John H. Wiggins 

SAC Project Management Committee 

SEAOC: William T. Holmes Program Manager: Stephen A. Mahin 

ATC: Christopher Rojahn 

CUREe: Robin Shepherd 

Project Director for Topical Investigations: 

James O. Malley 

Project Director for Product Development: 

Ronald O. Hamburger 

SAC Joint Venture 

SEAOC: www.seaoc.org 

ATC: www.atcouncil.org 

CUREe: www.curee.org 

June, 2000 

THE SAC JOINT VENTURE 

SAC is a joint venture of the Structural Engineers Association of California (SEAOC), the Applied 

Technology Council (ATC), and California Universities for Research in Earthquake Engineering 

(CUREe), formed specifically to address both immediate and long-term needs related to solving 

performance problems with welded, steel moment-frame connections discovered following the 1994 

Northridge earthquake. SEAOC is a professional organization composed of more than 3,000 practicing 

structural engineers in California. The volunteer efforts of SEAOC’s members on various technical 

committees have been instrumental in the development of the earthquake design provisions contained in 

the Uniform Building Code and the 1997 National Earthquake Hazards Reduction Program (NEHRP) 

Recommended Provisions for Seismic Regulations for New Buildings and Other Structures. ATC is a 

nonprofit corporation founded to develop structural engineering resources and applications to mitigate 

the effects of natural and other hazards on the built environment. Since its inception in the early 1970s, 

ATC has developed the technical basis for the current model national seismic design codes for buildings; 

the de-facto national standard for postearthquake safety evaluation of buildings; nationally applicable 

guidelines and procedures for the identification, evaluation, and rehabilitation of seismically hazardous 

buildings; and other widely used procedures and data to improve structural engineering practice. CUREe 

is a nonprofit organization formed to promote and conduct research and educational activities related to 

earthquake hazard mitigation. CUREe’s eight institutional members are the California Institute of 

Technology, Stanford University, the University of California at Berkeley, the University of California at 

Davis, the University of California at Irvine, the University of California at Los Angeles, the University 

of California at San Diego, and the University of Southern California. These university earthquake 

research laboratory, library, computer and faculty resources are among the most extensive in the United 

States. The SAC Joint Venture allows these three organizations to combine their extensive and unique 

resources, augmented by consultants and subcontractor universities and organizations from across the 

nation, into an integrated team of practitioners and researchers, uniquely qualified to solve problems 

related to the seismic performance of steel moment-frame structures. 

ACKNOWLEDGEMENTS 

Funding for Phases I and II of the SAC Steel Program to Reduce the Earthquake Hazards of Steel 

Moment-Frame Structures was principally provided by the Federal Emergency Management Agency, 

with ten percent of the Phase I program funded by the State of California, Office of Emergency Services. 

Substantial additional support, in the form of donated materials, services, and data has been provided by 

a number of individual consulting engineers, inspectors, researchers, fabricators, materials suppliers and 

industry groups. Special efforts have been made to maintain a liaison with the engineering profession, 

researchers, the steel industry, fabricators, code-writing organizations and model code groups, building 

officials, insurance and risk-management groups, and federal and state agencies active in earthquake 

hazard mitigation efforts. SAC wishes to acknowledge the support and participation of each of the above 

groups, organizations and individuals. In particular, we wish to acknowledge the contributions provided 

by the American Institute of Steel Construction, the Lincoln Electric Company, the National Institute of 

Standards and Technology, the National Science Foundation, and the Structural Shape Producers 

Council. SAC also takes this opportunity to acknowledge the efforts of the project participants – the 

managers, investigators, writers, and editorial and production staff – whose work has contributed to the 

development of these documents. Finally, SAC extends special acknowledgement to Mr. Michael 

Mahoney, FEMA Project Officer, and Dr. Robert Hanson, FEMA Technical Advisor, for their continued 

support and contribution to the success of this effort. 

Recommended Seismic Design 

Criteria For New Steel FEMA-350 

Moment-Frame Buildings Table of Contents 

TABLE OF CONTENTS 

LIST OF FIGURES ..............................................................................................................................ix

LIST OF TABLES................................................................................................................................xi

1 INTRODUCTION............................................................................................................ 1-1

1.1 Purpose................................................................................................................. 1-1

1.2 Intent .................................................................................................................... 1-2

1.3 Background .......................................................................................................... 1-3

1.4 Application......................................................................................................... 1-10

1.5 Overview............................................................................................................ 1-10

2 GENERAL REQUIREMENTS ....................................................................................... 2-1

2.1 Scope.................................................................................................................... 2-1

2.2 Applicable Codes, Standards, and References..................................................... 2-1

2.3 Basic Design Approach........................................................................................ 2-2

2.4 Design Performance Objectives........................................................................... 2-3

2.5 System Selection.................................................................................................. 2-7

2.5.1 Configuration and Load Path ................................................................... 2-7

2.5.2 Structural System Selection ..................................................................... 2-7

2.5.3 Connection Type ...................................................................................... 2-8

2.5.4 Redundancy.............................................................................................. 2-9

2.5.5 Frame Beam Spans................................................................................. 2-10

2.6 Structural Materials............................................................................................ 2-11

2.6.1 Material Specifications .......................................................................... 2-11

2.6.2 Material Strength Properties .................................................................. 2-12

2.7 Structural Analysis............................................................................................. 2-13

2.8 Mathematical Modeling..................................................................................... 2-13

2.8.1 Basic Assumptions................................................................................. 2-13

2.8.2 Model Configuration.............................................................................. 2-14

2.8.2.1 Regularity................................................................................ 2-14

2.8.2.2 Elements Modeled .................................................................. 2-14

2.1.1.3 Connection Stiffness ............................................................... 2-15

2.1.3 Horizontal Torsion................................................................................. 2-16

2.1.4 Foundation Modeling............................................................................. 2-16

2.1.5 Diaphragms ............................................................................................ 2-17

2.1.6 P-D Effects............................................................................................. 2-17

2.1.7 Multidirectional Excitation Effects........................................................ 2-20

2.1.8 Vertical Excitation ................................................................................. 2-20

2.9 Frame Design ................................................................................................. 2-21

2.9.1 Strength of Beams and Columns............................................................ 2-21

2.1.2 Lateral Bracing of Column Flanges ....................................................... 2-22

2.1.3 Panel Zone Strength............................................................................... 2-22

2.1.4 Section Compactness Requirements...................................................... 2-23

2.1.5 Beam Lateral Bracing ............................................................................ 2-23

iii 
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FEMA-350 Criteria for New Steel 

Table of Contents Moment-Frame Buildings 

2.1.6 Deep Columns........................................................................................ 2-23

2.1.7 Built-up Sections.................................................................................... 2-24

2.2 Connection Design............................................................................................. 2-24

2.3 Specifications..................................................................................................... 2-25

2.4 Quality Control and Quality Assurance ............................................................. 2-26

2.5 Other Structural Connections............................................................................. 2-26

2.5.1 Column Splices ...................................................................................... 2-26

2.5.2 Column Bases ........................................................................................ 2-28

2.5.3 Welded Collectors and Chords .............................................................. 2-29

2.5.4 Simple Beam-to-Column Gravity Connections ..................................... 2-29

CONNECTION QUALIFICATION................................................................................ 3-1

3.1 Scope.................................................................................................................... 3-1

3.2 Basic Design Approach........................................................................................ 3-2

3.2.1 Frame Configuration................................................................................ 3-2

3.2.2 Connection Configuration........................................................................ 3-5

3.2.3 Determine Plastic Hinge Locations.......................................................... 3-5

3.2.4 Determine Probable Plastic Moment at Hinges ....................................... 3-6

3.2.5 Determine Shear at the Plastic Hinge ...................................................... 3-7

3.2.6 Determine Strength Demands at Each Critical Section ........................... 3-7

3.2.7 Yield Moment .......................................................................................... 3-8

3.3 General Requirements.......................................................................................... 3-9

3.3.1 Beams....................................................................................................... 3-9

3.3.1.1 Beam Flange Stability......................................................... 3-9

3.3.1.2 Beam Web Stability .......................................................... 3-10

3.3.1.3 Beam Depth and Span Effects .......................................... 3-10

3.3.1.4 Beam Flange Thickness Effects........................................ 3-11

3.3.1.5 Lateral Bracing at Beam Flanges at Plastic Hinges .......... 3-11

3.3.1.6 Welded Shear Studs .......................................................... 3-12

3.3.2 Welded Joints......................................................................................... 3-12

3.3.2.1 Through-Thickness Strength............................................. 3-12

3.3.2.2 Base Material Toughness.................................................. 3-13

3.3.2.3 k-Area Properties .............................................................. 3-14

3.3.2.4 Weld Metal Matching and Overmatching......................... 3-15

3.3.2.5 Weld Metal Toughness ..................................................... 3-15

3.3.2.6 Weld Backing, Weld Tabs and Other Details................... 3-16

3.3.2.7 Weld Access Holes ........................................................... 3-17

3.3.2.8 Welding Quality Control and Quality Assurance ............. 3-17

3.3.3 Other Design Issues for Welded Connections ....................................... 3-19

3.3.3.1 Continuity Plates............................................................... 3-19

3.3.3.2 Panel Zone Strength.......................................................... 3-21

3.3.3.3 Connections to Column Minor Axis................................. 3-22

3.3.3.4 Attachment of Other Construction.................................... 3-22

3.3.4 Bolted Joints .......................................................................................... 3-23

3.4 Prequalified Connections – General .................................................................. 3-23

iv 

3 
Recommended Seismic Design 

Criteria For New Steel FEMA-350 

Moment-Frame Buildings Table of Contents 

3.4.1 Load Combinations and Resistance Factors .......................................... 3-25

3.5 Prequalified Welded Fully Restrained Connections .......................................... 3-25

3.5.1 Welded Unreinforced Flange – Bolted Web Connections..................... 3-26

3.5.1.1 Design Procedure .............................................................. 3-28

3.5.2 Welded Unreinforced Flange – Welded Web Connections ................... 3-28

3.5.2.1 Design Procedure .............................................................. 3-31

3.5.3 Free Flange Connections........................................................................ 3-31

3.5.3.1 Design Procedure ............................................................. 3-33

3.5.4 Welded Flange Plate Connections ......................................................... 3-34

3.5.4.1 Design Procedure .............................................................. 3-37

3.5.5 Reduced Beam Section Connections ..................................................... 3-38

3.5.5.1 Design Procedure .............................................................. 3-41

3.5.5.2 Fabrication Requirements ................................................. 3-42

3.5.5.3 Composite Construction ................................................... 3-42

3.6 Prequalified Bolted Fully Restrained Connections............................................ 3-42

3.6.1 Bolted Unstiffened End Plate Connections............................................ 3-42

3.6.1.1 Design Procedure .............................................................. 3-45

3.6.2 Bolted Stiffened End Plate Connection ................................................. 3-48

3.6.2.1 Design Procedure .............................................................. 3-51

3.6.3 Bolted Flange Plate Connections ........................................................... 3-53

3.6.3.1 Design Procedure .............................................................. 3-56

3.7 Prequalified Partially Restrained Connections .................................................. 3-59

3.7.1 Double Split Tee Connections ............................................................... 3-60

3.7.1.1 Connection Stiffness ......................................................... 3-63

3.7.1.2 Design Procedure .............................................................. 3-63

3.8 Proprietary Connections..................................................................................... 3-68

3.8.1 Side Plate ............................................................................................... 3-68

3.8.2 Slotted Web............................................................................................ 3-70

3.8.3 Bolted Bracket........................................................................................ 3-72

3.8.4 Reduced Web......................................................................................... 3-72

3.9 Project-Specific Connection Qualification ........................................................ 3-73

3.9.1 Testing Procedures................................................................................. 3-74

3.9.2 Acceptance Criteria................................................................................ 3-76

3.9.3 Analytical Prediction of Behavior.......................................................... 3-78

3.10 Prequalification Testing Criteria........................................................................ 3-78

3.10.1 Prequalification Testing......................................................................... 3-79

3.10.2 Extending the Limits on Prequalified Connections ............................... 3-79

4 PERFORMANCE EVALUATION ................................................................................. 4-1

4.1 Scope.................................................................................................................... 4-1

4.2 Performance Definition........................................................................................ 4-1

4.2.1 Hazard Specification................................................................................ 4-3

4.2.1.1 General ................................................................................. 4-3

4.2.1.2 Ground Shaking.................................................................... 4-3

4.2.1.3 Other Hazards....................................................................... 4-5

v 
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Table of Contents Moment-Frame Buildings 

4.2.2 Performance Levels.................................................................................. 4-5

4.2.2.1 Nonstructural Performance Levels ....................................... 4-7

4.2.2.2 Structural Performance Levels ............................................. 4-7

4.2.2.2.1 Collapse Prevention Performance Level................. 4-8

4.2.2.2.2 Immediate Occupancy Performance Level ............. 4-8

4.3 Evaluation Approach ......................................................................................... 4-10

4.4 Analysis.............................................................................................................. 4-11

4.4.1 Alternative Procedures........................................................................... 4-11

4.4.2 Procedure Selection .............................................................................. 4-13

4.4.3 Linear Static Procedure.......................................................................... 4-13

4.4.3.1 Basis of the Procedure....................................................... 4-13

4.4.3.2 Period Determination........................................................ 4-15

4.4.3.3 Determination of Actions and Deformations.................... 4-16

4.4.3.3.1 Psuedo Lateral Load.............................................. 4-16

4.4.3.3.2 Vertical Distribution of Seismic Forces................ 4-18

4.4.3.3.3 Horizontal Distribution of Seismic Forces ........... 4-18

4.4.3.3.4 Diaphragms ........................................................... 4-18

4.4.3.3.5 Determination of Interstory Drift .......................... 4-18

4.4.3.3.6 Determination of Column Demands ..................... 4-19

4.4.4 Linear Dynamic Procedure..................................................................... 4-19

4.4.4.1 Basis of the Procedure....................................................... 4-19

4.4.4.2 Analysis............................................................................. 4-20

4.4.4.2.1 General.................................................................. 4-20

4.4.4.2.2 Ground Motion Characterization .......................... 4-21

4.4.4.3 Determination of Actions and Deformations.................... 4-21

4.4.4.3.1 Factored Interstory Drift Demand......................... 4-21

4.4.4.3.2 Determination of Column Demands ..................... 4-21

4.4.5 Nonlinear Static Procedure .................................................................... 4-21

4.4.5.1 Basis of the Procedure....................................................... 4-21

4.4.5.2 Analysis Considerations.................................................... 4-22

4.4.5.2.1 General.................................................................. 4-22

4.4.5.2.2 Control Node......................................................... 4-23

4.4.5.2.3 Lateral Load Patterns ............................................ 4-24

4.4.5.2.4 Period Determination............................................ 4-24

4.4.5.2.5 Analysis of Three-Dimensional Models ............... 4-24

4.4.5.2.6 Analysis of Two-Dimensional Models ................. 4-24

4.4.5.3 Determination of Actions and Deformations.................... 4-24

4.4.5.3.1 Target Displacement ............................................. 4-24

4.4.5.3.2 Diaphragms ........................................................... 4-24

4.4.5.3.3 Factored Interstory Drift Demand......................... 4-24

4.4.5.3.4 Multidirectional effects......................................... 4-25

4.4.5.3.5 Factored Column and Column Splice Demands... 4-25

4.4.6 Nonlinear Dynamic Procedure............................................................... 4-25

4.4.6.1 Basis of the Procedure....................................................... 4-25

vi 
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Criteria For New Steel FEMA-350 

Moment-Frame Buildings Table of Contents 

4.4.6.2 Analysis Assumptions....................................................... 4-25

4.4.6.2.1 General.................................................................. 4-25

4.4.6.2.2 Ground Motion Characterization .......................... 4-26

4.4.6.3 Determination of Actions and Deformations.................... 4-26

4.4.6.3.1 Response Quantities.............................................. 4-26

4.4.6.3.2 Factored Interstory Drift Demand......................... 4-26

4.4.6.3.3 Factored Column and Column Splice Demands... 4-26

4.5 Mathematical Modeling ............................................................................... 4-26

4.5.1 Basic Assumptions ............................................................................... 4-26

4.5.2 Frame Configuration.............................................................................. 4-27

4.5.2.1 Modeling........................................................................... 4-27

4.5.2.2 Connection Modeling........................................................ 4-28

4.5.2.2.1 Fully Restrained Moment-Resisting

Connections........................................................... 4-28

4.5.2.2.2 Partially Restrained Moment-Resisting

Connections........................................................... 4-28

4.5.2.2.3 Simple Shear Tab Connections............................. 4-29

4.5.2.3 Panel Zone Stiffness ......................................................... 4-29

4.5.3 Horizontal Torsion ............................................................................... 4-30

4.5.4 Foundation Modeling............................................................................. 4-31

4.5.5 Diaphragms ........................................................................................... 4-31

4.5.6 P-D Effects ........................................................................................... 4-32

4.5.7 Multidirectional Excitation Effects........................................................ 4-32

4.5.8 Vertical Ground Motion......................................................................... 4-32

4.6 Acceptance Criteria ........................................................................................... 4-33

4.6.1 Factored-Demand-to Capacity Ratio...................................................... 4-33

4.6.2 Performance Limited By Interstory Drift Angle .................................... 4-36

4.6.2.1 Factored Interstory Drift Angle Demand .......................... 4-36

4.6.2.2 Factored Interstory Drift Angle Capacity.......................... 4-37

4.6.2.2.1 Global Interstory Drift Angle................................ 4-37

4.6.2.2.2 Local Interstory Drift Angle.................................. 4-37

4.6.3 Performance Limited by Column Compressive Capacity...................... 4-40

4.6.3.1 Column Compressive Demand ......................................... 4-40

4.6.3.2 Column Compressive Capacity......................................... 4-42

4.6.4 Column Splice Capacity......................................................................... 4-42

4.6.4.1 Column Splice Tensile Demand ....................................... 4-42

4.6.4.2 Column Splice Tensile Capacity....................................... 4-42

APPENDIX A: DETAILED PROCEDURE FOR PERFORMANCE EVALUATION ........... A-1

A.1 Scope................................................................................................................... A-1

A.2 Performance Evaluation Approach ..................................................................... A-1

A.2.1 Performance Objectives and Confidence................................................ A-1

A.2.2 Basic Procedure....................................................................................... A-4

A.3 Determination of Hazard Parameters.................................................................. A-7

A.3.1 Spectral Response Acceleration.............................................................. A-7

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Table of Contents Moment-Frame Buildings 

A.3.2 Logarithmic Hazard Curve Slope ........................................................... A-7

A.4 Determination of Demand Factors.................................................................... A-10

A.5 Determination of Beam-Column Connection Assembly Capacities................. A-13

A.5.1 Connection Test Protocols .................................................................... A-14

A.5.2 Determination of Beam-Column Assembly Capacities and

Resistance Factors................................................................................. A-14

A.6 Global Stability Capacity.................................................................................. A-15

REFERENCES, BIBLIOGRAPHY, AND ACRONYMS...........................................................R-1

SAC PROJECT PARTICIPANTS...............................................................................................S-1

viii 
Recommended Seismic Design

Criteria for New Steel FEMA-350

Moment-Frame Buildings List of Figures 

LIST OF FIGURES

Figure 1-1 Typical Welded Moment-Resisting Connection Prior to 1994............................ 1-4

Figure 1-2 Common Zone of Fracture Initiation in Beam-Column Connection ................... 1-5

Figure 1-3 Fractures of Beam-to-Column Joints ................................................................... 1-5

Figure 1-4 Column Fractures................................................................................................. 1-6

Figure 1-5 Vertical Fracture through Beam Shear Plate Connection .................................... 1-6

Figure 2-1 NEHRP Seismic Use Groups (SUG) and Performance ....................................... 2-4

Figure 2-2 Interstory Drift Angle........................................................................................... 2-9

Figure 3-1 Inelastic Behavior of Frames with Hinges in Beam Span.................................... 3-3

Figure 3-2 Location of Plastic Hinge Formation ................................................................... 3-6

Figure 3-3 Sample Calculation of Shear at Plastic Hinge ..................................................... 3-8

Figure 3-4 Calculation of Demands at Critical Sections ....................................................... 3-8

Figure 3-5 Recommended Weld Access Hole Detail .......................................................... 3-18

Figure 3-6 Typical Continuity and Doubler Plates.............................................................. 3-20

Figure 3-7 Welded Unreinforced Flange – Bolted Web (WUF-B) Connection.................. 3-26

Figure 3-8 Welded Unreinforced Flange – Welded Web (WUF-W) Connection............... 3-29

Figure 3-9 Welded Free Flange (FF) Connection................................................................ 3-33

Figure 3-10 Schematic of the Forces for Design of the Free Flange Shear Tab.................... 3-35

Figure 3-11 Welded Flange Plate (WFP) Connection........................................................... 3-36

Figure 3-12 Reduced Beam Section (RBS) Connection........................................................ 3-39

Figure 3-13 Bolted Unstiffened End Plate (BUEP) Connection ........................................... 3-43

Figure 3-14 Geometry of Unstiffened End Plate Connection................................................ 3-48

Figure 3-15 Stiffened End Plate Connection......................................................................... 3-49

Figure 3-16 Geometry of Stiffened End Plate Connection.................................................... 3-53

Figure 3-17 Bolted Flange Plate (BFP) Connection.............................................................. 3-54

Figure 3-18 Geometry of the Bolted Flange Plate Connection.............................................. 3-56

Figure 3-19 Block Shear and Pull-Through Failures............................................................. 3-59

Figure 3-20 Double Split Tee (DST) Connection.................................................................. 3-61

Figure 3-21 Geometry for Prying Forces and Bending of T-Section Flanges ....................... 3-64

Figure 3-22 Geometry for Other T-Stub Failure Modes........................................................ 3-64

Figure 3-23 Proprietary Side Plate Connection ..................................................................... 3-69

Figure 3-24 Proprietary Slotted Web Connection ................................................................. 3-71

Figure 3-25 Bolted Bracket Connection ................................................................................ 3-72

Figure 3-26 Reduced Web Connection.................................................................................. 3-73

Figure 3-27 Angular Rotation of Test Assembly................................................................... 3-74

Figure A-1 Representative Incremental Dynamic Analysis Plots ....................................... A-17

ix 

Recommended Seismic Design

Moment-Frame Buildings List of Tables

Criteria for New Steel FEMA-350 

LIST OF TABLES

Table 2-1 Values of Ry for Various Material Grades ......................................................... 2-12

Table 2-2 Prequalified Connection Details ........................................................................ 2-25

Table 3-1 Prequalified Welded Fully Restrained Connections .......................................... 3-25

Table 3-2 Prequalification Data WUF-B Connections....................................................... 3-27

Table 3-3 Prequalification Data WUF-W Connections ..................................................... 3-30

Table 3-4 Prequalification Data for Free Flange Connections........................................... 3-32

Table 3-5 Prequalification Data for WFP Connections...................................................... 3-35

Table 3-6 Prequalification Data for RBS Connections ...................................................... 3-40

Table 3-7 Prequalified Bolted Fully Restrained Connections............................................ 3-42

Table 3-8 Prequalification Data for BUEP Connections.................................................... 3-44

Table 3-9 Prequalification Data for Bolted Stiffened End Plate Connections ................... 3-50

Table 3-10 Prequalification Data for Bolted Flange Plate Connections .............................. 3-55

Table 3-11 Prequalified Bolted Partially Restrained Connections....................................... 3-60

Table 3-12 Prequalification Data for Full Strength DST Connections (FSDST)................. 3-62

Table 3-13 Interstory Drift Angle Limits for Various Performance Levels ......................... 3-74

Table 3-14 Numerical Values of q j and nj ........................................................................... 3-75

Table 3-15 Minimum Qualifying Total Interstory Drift Angle Capacities, qSD, and qU

for OMF and SMF Systems ............................................................................... 3-76

Table 4-1 Building Performance Levels............................................................................... 4-6

Table 4-2 Structural Performance Levels............................................................................. 4-8

Table 4-3 Analysis Procedure Selection Criteria ............................................................... 4-14

Table 4-4 Modification Factor C3 for the Linear Static Procedure .................................... 4-18

Table 4-5 Performance Parameters Requiring Evaluation of Confidence ......................... 4-33

Table 4-6 Confidence Levels for Various Values of l, Given bUT .................................... 4-35

Table 4-7 Recommended Minimum Confidence Levels.................................................... 4-35

Table 4-8 Interstory Drift Angle Analysis Uncertainty Factors ga ..................................... 4-36

Table 4-9 Interstory Drift Angle Demand Variability Factors g......................................... 4-37

Table 4-10 Global Interstory Drift Angle Capacity C and Resistance Factors f for

Regular SMF and OMF Buildings ..................................................................... 4-38

Table 4-11 Uncertainty Coefficient bUT for Global Interstory Drift Evaluation .................. 4-38

Table 4-12 Drift Angle Capacity C(q10, qU) for Prequalified Connections as

Limited by Local Connection Response ............................................................ 4-39

Table 4-13 Uncertainty Coefficient bUT for Local Interstory Drift Evaluation .................... 4-40

Table 4-14 Behavior States for Performance Evaluation of Connection Assemblies.......... 4-40

Table 4-15 Analysis Uncertainty Factor ga and Total Uncertainty Coefficient bUT for

Evaluation of Column Compressive Demands.................................................. 4-41

Table A-1 Confidence Parameter, l, as a Function of Confidence Level,

Hazard Parameter k, and Uncertainty bUT ........................................................... A-8

Table A-2 Default Values of the Logarithmic Hazard Curve Slope k for

Probabilistic Ground Shaking Hazards............................................................... A-9

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Table of Contents Moment-Frame Buildings 

Table A-3 Default Logarithmic Uncertainty bDU for Various Analysis Methods.............. A-12

Table A-4 Default Bias Factors CB .................................................................................... A-12

Table A-5 Behavior States for Performance Evaluation of Connection Assemblies......... A-14

xii 
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Moment-Frame Buildings Chapter 1: Introduction

Criteria for New Steel FEMA-350 

1. INTRODUCTION 

1.1 Purpose 

This report, FEMA-350 – Recommended Seismic Design Criteria for New Steel MomentFrame Buildings has been developed by the SAC Joint Venture under contract to the Federal 

Emergency Management Agency (FEMA) to provide organizations engaged in the development 

of consensus design standards and building code provisions with recommended criteria for the 

design and construction of new buildings incorporating moment-resisting steel frame 

construction to resist the effects of earthquakes. It is one of a series of companion publications 

addressing the issue of the seismic performance of steel moment-frame buildings. The set of 

companion publications includes: 

• FEMA-350 – Recommended Seismic Design Criteria for New Steel Moment-Frame 

Buildings. This publication provides recommended criteria, supplemental to FEMA-302 – 

1997 NEHRP Recommended Provisions for Seismic Regulations for New Buildings and 

Other Structures, for the design and construction of steel moment-frame buildings and 

provides alternative performance-based design criteria. 

• FEMA-351 – Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded 

Steel Moment-Frame Buildings. This publication provides recommended methods to 

evaluate the probable performance of existing steel moment-frame buildings in future 

earthquakes and to retrofit these buildings for improved performance. 

• FEMA-352 – Recommended Postearthquake Evaluation and Repair Criteria for Welded 

Steel Moment-Frame Buildings. This publication provides recommendations for performing 

postearthquake inspections to detect damage in steel moment-frame buildings following an 

earthquake, evaluating the damaged buildings to determine their safety in the postearthquake 

environment, and repairing damaged buildings. 

• FEMA-353 – Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications. This publication provides 

recommended specifications for the fabrication and erection of steel moment frames for 

seismic applications. The recommended design criteria contained in the other companion 

documents are based on the material and workmanship standards contained in this document, 

which also includes discussion of the basis for the quality control and quality assurance 

criteria contained in the recommended specifications. 

The information contained in these recommended design criteria, hereinafter referred to as 

Recommended Criteria, is presented in the form of specific design and performance evaluation 

procedures together with supporting commentary explaining part of the basis for these 

recommendations. Detailed derivations and explanations of the basis for these design and 

evaluation recommendations may be found in a series of State of the Art Reports prepared in 

parallel with these Recommended Criteria. These reports include: 

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Chapter 1: Introduction Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

• FEMA-355A – State of the Art Report on Base Metals and Fracture. This report summarizes 

current knowledge of the properties of structural steels commonly employed in building 

construction, and the production and service factors that affect these properties. 

• FEMA-355B – State of the Art Report on Welding and Inspection. This report summarizes 

current knowledge of the properties of structural welding commonly employed in building 

construction, the effect of various welding parameters on these properties, and the 

effectiveness of various inspection methodologies in characterizing the quality of welded 

construction. 

• FEMA-355C – State of the Art Report on Systems Performance of Steel Moment Frames 

Subject to Earthquake Ground Shaking. This report summarizes an extensive series of 

analytical investigations into the demands induced in steel moment-frame buildings designed 

to various criteria, when subjected to a range of different ground motions. The behavior of 

frames constructed with fully restrained, partially restrained and fracture-vulnerable 

connections is explored for a series of ground motions, including motion anticipated at nearfault and soft-soil sites. 

• FEMA-355D – State of the Art Report on Connection Performance. This report summarizes 

the current state of knowledge of the performance of different types of moment-resisting 

connections under large inelastic deformation demands. It includes information on fully 

restrained, partially restrained, and partial strength connections, both welded and bolted, 

based on laboratory and analytical investigations. 

• FEMA-355E – State of the Art Report on Past Performance of Steel Moment-Frame 

Buildings in Earthquakes. This report summarizes investigations of the performance of steel 

moment-frame buildings in past earthquakes, including the 1995 Kobe, 1994 Northridge, 

1992 Landers, 1992 Big Bear, 1989 Loma Prieta and 1971 San Fernando events. 

• FEMA-355F – State of the Art Report on Performance Prediction and Evaluation of Steel 

Moment-Frame Buildings. This report describes the results of investigations into the ability 

of various analytical techniques, commonly used in design, to predict the performance of 

steel moment-frame buildings subjected to earthquake ground motion. Also presented is the 

basis for performance-based evaluation procedures contained in the design criteria 

documents, FEMA-350, FEMA-351, and FEMA-352. 

In addition to the recommended design criteria and the State of the Art Reports, a companion 

document has been prepared for building owners, local community officials and other nontechnical audiences who need to understand this issue. A Policy Guide to Steel Moment Frame 

Construction (FEMA-354) addresses the social, economic, and political issues related to the 

earthquake performance of steel moment-frame buildings. FEMA-354 also includes discussion 

of the relative costs and benefits of implementing the recommended criteria. 

1.2 Intent 

These Recommended Criteria are primarily intended as a resource document for organizations 

engaged in the development of building codes and consensus standards for regulation of the design 

and construction of steel moment-frame structures that may be subject to the effects of earthquake 

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ground shaking. These criteria have been developed by professional engineers and researchers, 

based on the findings of a large multi-year program of investigation and research into the 

performance of steel moment-frame structures. Development of these recommended criteria was 

not subjected to a formal consensus review and approval process, nor was formal review or 

approval obtained from SEAOC’s technical committees. However, it did include broad external 

review by practicing engineers, researchers, fabricators, and the producers of steel and welding 

consumables. In addition, two workshops were convened to obtain direct comment from these 

stakeholders on the proposed recommendations. 

1.3 Background 

For many years, the basic intent of the building code seismic provisions has been to provide 

buildings with an ability to withstand intense ground shaking without collapse, but potentially 

with some significant structural damage. In order to accomplish this, one of the basic principles 

inherent in modern code provisions is to encourage the use of building configurations, structural 

systems, materials and details that are capable of ductile behavior. A structure is said to behave 

in a ductile manner if it is capable of withstanding large inelastic deformations without 

significant degradation in strength, and without the development of instability and collapse. The 

design forces specified by building codes for particular structural systems are related to the 

amount of ductility the system is deemed to possess. Generally, structural systems with more 

ductility are designed for lower forces than less ductile systems, as ductile systems are deemed 

capable of resisting demands that are significantly greater than their elastic strength limit. 

Starting in the 1960s, engineers began to regard welded steel moment-frame buildings as being 

among the most ductile systems contained in the building code. Many engineers believed that 

steel moment-frame buildings were essentially invulnerable to earthquake-induced structural 

damage and thought that should such damage occur, it would be limited to ductile yielding of 

members and connections. Earthquake-induced collapse was not believed possible. Partly as a 

result of this belief, many large industrial, commercial and institutional structures employing 

steel moment-frame systems were constructed, particularly in the western United States. 

The Northridge earthquake of January 17, 1994 challenged this paradigm. Following that 

earthquake, a number of steel moment-frame buildings were found to have experienced brittle 

fractures of beam-to-column connections. The damaged buildings had heights ranging from one 

story to 26 stories, and a range of ages spanning from buildings as old as 30 years to structures 

being erected at the time of the earthquake. The damaged buildings were spread over a large 

geographical area, including sites that experienced only moderate levels of ground shaking. 

Although relatively few buildings were located on sites that experienced the strongest ground 

shaking, damage to buildings on these sites was extensive. Discovery of these unanticipated 

brittle fractures of framing connections, often with little associated architectural damage, was 

alarming to engineers and the building industry. The discovery also caused some concern that 

similar, but undiscovered, damage may have occurred in other buildings affected by past 

earthquakes. Later investigations confirmed such damage in a limited number of buildings 

affected by the 1992 Landers, 1992 Big Bear and 1989 Loma Prieta earthquakes. 

In general, steel moment-frame buildings damaged by the Northridge earthquake met the 

basic intent of the building codes. That is, they experienced limited structural damage, but did 

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not collapse. However, the structures did not behave as anticipated and significant economic 

losses occurred as a result of the connection damage, in some cases, in buildings that had 

experienced ground shaking less severe than the design level. These losses included direct costs 

associated with the investigation and repair of this damage as well as indirect losses relating to 

the temporary, and in a few cases, long-term, loss of use of space within damaged buildings. 

Steel moment-frame buildings are designed to resist earthquake ground shaking based on the 

assumption that they are capable of extensive yielding and plastic deformation, without loss of 

strength. The intended plastic deformation consists of plastic rotations developing within the 

beams, at their connections to the columns, and is theoretically capable of resulting in benign 

dissipation of the earthquake energy delivered to the building. Damage is expected to consist of 

moderate yielding and localized buckling of the steel elements, not brittle fractures. Based on this 

presumed behavior, building codes permit steel moment-frame buildings to be designed with a 

fraction of the strength that would be required to respond to design level earthquake ground shaking 

in an elastic manner. 

Steel moment-frame buildings are anticipated to develop their ductility through the 

development of yielding in beam-column assemblies at the beam-column connections. This 

yielding may take the form of plastic hinging in the beams (or, less desirably, in the columns), 

plastic shear deformation in the column panel zones, or through a combination of these 

mechanisms. It was believed that the typical connection employed in steel moment-frame 

construction, shown in Figure 1-1, was capable of developing large plastic rotations, on the order 

of 0.02 radians or larger, without significant strength degradation. 

Figure 1-1 Typical Welded Moment-Resisting Connection Prior to 1994 

Observation of damage sustained by buildings in the 1994 Northridge earthquake indicated 

that, contrary to the intended behavior, in many cases, brittle fractures initiated within the 

connections at very low levels of plastic demand, and in some cases, while the structures 

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Backing bar Backing bar

Beam flange Beam flange

Fracture Fracture

Backing bar

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Moment-Frame Buildings Chapter 1: Introduction

Criteria for New Steel FEMA-350 

remained essentially elastic. Typically, but not always, fractures initiated at the complete joint 

penetration (CJP) weld between the beam bottom flange and column flange (Figure 1-2). Once 

initiated, these fractures progressed along a number of different paths, depending on the 

individual joint conditions. 

Column flange Column flange

Fused zone Fused zone

Beam flange

Backing bar

Fracture

Figure 1-2 Common Zone of Fracture Initiation in Beam -Column Connection 

In some cases, the fractures progressed completely through the thickness of the weld, and 

when fire protective finishes were removed, the fractures were evident as a crack through 

exposed faces of the weld, or the metal just behind the weld (Figure 1-3a). Other fracture 

patterns also developed. In some cases, the fracture developed into a crack of the column flange 

material behind the CJP weld (Figure 1-3b). In these cases, a portion of the column flange 

remained bonded to the beam flange, but pulled free from the remainder of the column. This 

fracture pattern has sometimes been termed a “divot” or “nugget” failure. 

A number of fractures progressed completely through the column flange, along a nearhorizontal plane that aligns approximately with the beam lower flange (Figure 1-4a). In some 

cases, these fractures extended into the column web and progressed across the panel zone (Figure 

1-4b). Investigators have reported some instances where columns fractured entirely across the 

section. 

a. Fracture at Fused Zone b. Column Flange "Divot" Fracture 

Figure 1-3 Fractures of Beam-to-Column Joints 

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a. Fractures through Column Flange b. Fracture Progresses into Column Web 

Figure 1-4 Column Fractures 

Once such fractures have occurred, the beam-column connection has experienced a 

significant loss of flexural rigidity and strength to resist those loads that tend to open the crack. 

Residual flexural strength and rigidity must be developed through a couple consisting of forces 

transmitted through the remaining top flange connection and the web bolts. However, in 

providing this residual strength and stiffness, the bolted web connections can themselves be 

subject to failures. These include fracturing of the welds of the shear plate to the column, 

fracturing of supplemental welds to the beam web or fracturing through the weak section of 

shear plate aligning with the bolt holes (Figure 1-5). 

Despite the obvious local strength impairment resulting from these fractures, many damaged 

buildings did not display overt signs of structural damage, such as permanent drifts or damage to 

architectural elements, making reliable postearthquake damage evaluations difficult. In order to 

determine if a building has sustained connection damage it is necessary to remove architectural 

finishes and fireproofing, and perform detailed inspections of the connections. Even if no 

damage is found, this is a costly process. Repair of damaged connections is even more costly. 

At least one steel moment-frame building sustained so much damage that it was deemed more 

practical to demolish the building than to repair it. 

Figure 1-5 Vertical Fracture through Beam Shear Plate Connection 

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Initially, the steel construction industry took the lead in investigating the causes of this 

unanticipated damage and in developing design recommendations. The American Institute of 

Steel Construction (AISC) convened a special task committee in March, 1994 to collect and 

disseminate available information on the extent of the problem (AISC, 1994a). In addition, 

together with a private party engaged in the construction of a major steel building at the time of 

the earthquake, AISC participated in sponsoring a limited series of tests of alternative connection 

details at the University of Texas at Austin (AISC, 1994b). The American Welding Society 

(AWS) also convened a special task group to investigate the extent to which the damage was 

related to welding practice, and to determine if changes to the welding code were appropriate 

(AWS, 1995). 

In September, 1994, the SAC Joint Venture, AISC, the American Iron and Steel Institute and 

National Institute of Standards and Technology jointly convened an international workshop 

(SAC, 1994) in Los Angeles to coordinate the efforts of the various participants and to lay the 

foundation for systematic investigation and resolution of the problem. Following this workshop, 

FEMA entered into a cooperative agreement with the SAC Joint Venture to perform problemfocused studies of the seismic performance of steel moment-frame buildings and to develop 

recommendations for professional practice (Phase I of SAC Steel Project). Specifically, these 

recommendations were intended to address the following: the inspection of earthquake-affected 

buildings to determine if they had sustained significant damage; the repair of damaged buildings; 

the upgrade of existing buildings to improve their probable future performance; and the design of 

new structures to provide reliable seismic performance. 

During the first half of 1995, an intensive program of research was conducted to explore 

more definitively the pertinent issues. This research included literature surveys, data collection 

on affected structures, statistical evaluation of the collected data, analytical studies of damaged 

and undamaged buildings, and laboratory testing of a series of full-scale beam-column 

assemblies representing typical pre-Northridge design and construction practice as well as 

various repair, upgrade and alternative design details. The findings of these tasks formed the 

basis for the development of FEMA-267 – Interim Guidelines: Evaluation, Repair, Modification, 

and Design of Welded Steel Moment Frame Structures, which was published in August, 1995. 

FEMA-267 provided the first definitive, albeit interim, recommendations for practice, following 

the discovery of connection damage in the 1994 Northridge earthquake. 

In September 1995 the SAC Joint Venture entered into a contractual agreement with FEMA 

to conduct Phase II of the SAC Steel Project. Under Phase II, SAC continued its extensive 

problem-focused study of the performance of moment resisting steel frames and connections of 

various configurations, with the ultimate goal of develop seismic design criteria for steel 

construction. This work has included: extensive analyses of buildings; detailed finite element 

and fracture mechanics investigations of various connections to identify the effects of connection 

configuration, material strength, and toughness and weld joint quality on connection behavior; as 

well as more than 120 full-scale tests of connection assemblies. As a result of these studies, and 

independent research conducted by others, it is now known that the typical moment-resisting 

connection detail employed in steel moment-frame construction prior to the 1994 Northridge 

earthquake, and depicted in Figure 1-1, had a number of features that rendered it inherently 

susceptible to brittle fracture. These included the following: 

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• The most severe stresses in the connection assembly occur where the beam joins to the 

column. Unfortunately, this is also the weakest location in the assembly. At this location, 

bending moments and shear forces in the beam must be transferred to the column through the 

combined action of the welded joints between the beam flanges and column flanges and the 

shear tab. The combined section properties of these elements, for example the cross sectional 

area and section modulus, are typically less than those of the connected beam. As a result, 

stresses are locally intensified at this location. 

• The joint between the bottom beam flange and the column flange is typically made as a 

downhand field weld, often by a welder sitting on top of the beam top flange, in a so-called 

“wildcat” position. To make the weld from this position each pass must be interrupted at the 

beam web, with either a start or stop of the weld at this location. This welding technique 

often results in poor quality welding at this critical location, with slag inclusions, lack of 

fusion and other defects. These defects can serve as crack initiators, when the connection is 

subjected to severe stress and strain demands. 

• The basic configuration of the connection makes it difficult to detect hidden defects at the 

root of the welded beam-flange-to-column-flange joints. The backing bar, which was 

typically left in place following weld completion, restricts visual observation of the weld 

root. Therefore, the primary method of detecting defects in these joints is through the use of 

ultrasonic testing (UT). However, the geometry of the connection also makes it very difficult 

for UT to detect flaws reliably at the bottom beam flange weld root, particularly at the center 

of the joint, at the beam web. As a result, many of these welded joints have undetected 

significant defects that can serve as crack initiators. 

• Although typical design models for this connection assume that nearly all beam flexural 

stresses are transmitted by the flanges and all beam shear forces by the web, in reality, due to 

boundary conditions imposed by column deformations, the beam flanges at the connection 

carry a significant amount of the beam shear. This results in significant flexural stresses on 

the beam flange at the face of the column, and also induces large secondary stresses in the 

welded joint. Some of the earliest investigations of these stress concentration effects in the 

welded joint were conducted by Richard, et al. (1995). The stress concentrations resulting 

from this effect resulted in severe strength demands at the root of the complete joint 

penetration welds between the beam flanges and column flanges, a region that often includes 

significant discontinuities and slag inclusions, which are ready crack initiators. 

• In order that the welding of the beam flanges to the column flanges be continuous across the 

thickness of the beam web, this detail incorporates weld access holes in the beam web, at the 

beam flanges. Depending on their geometry, severe strain concentrations can occur in the 

beam flange at the toe of these weld access holes. These strain concentrations can result in 

low-cycle fatigue and the initiation of ductile tearing of the beam flanges after only a few 

cycles of moderate plastic deformation. Under large plastic flexural demands, these ductile 

tears can quickly become unstable and propagate across the beam flange. 

• Steel material at the center of the beam-flange-to-column-flange joint is restrained from 

movement, particularly in connections of heavy sections with thick column flanges. This 

condition of restraint inhibits the development of yielding at this location, resulting in locally 

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Moment-Frame Buildings Chapter 1: Introduction

Criteria for New Steel FEMA-350 

high stresses on the welded joint, which exacerbates the tendency to initiate fractures at 

defects in the welded joints. 

• Design practice in the period 1985-1994 encouraged design of these connections with 

relatively weak panel zones. In connections with excessively weak panel zones, inelastic 

behavior of the assembly is dominated by shear deformation of the panel zone. This panel 

zone shear deformation results in a local kinking of the column flanges adjacent to the beamflange-to-column-flange joint, and further increases the stress and strain demands in this 

sensitive region. 

In addition to the above, additional conditions contributed significantly to the vulnerability of 

connections constructed prior to 1994. 

• In the mid-1960s, the construction industry moved to the use of the semi-automatic, selfshielded, flux-cored arc welding process (FCAW-S) for making the joints of these 

connections. The welding consumables that building erectors most commonly used 

inherently produced welds with very low toughness. The toughness of this material could be 

further compromised by excessive deposition rates, which unfortunately were commonly 

employed by welders. As a result, brittle fractures could initiate in welds with large defects, 

at stresses approximating the yield strength of the beam steel, precluding the development of 

ductile behavior. 

• Early steel moment frames tended to be highly redundant and nearly every beam-column 

joint was constructed to behave as part of the lateral-force-resisting system. As a result, 

member sizes in these early frames were small and much of the early acceptance testing of 

this typical detail was conducted with specimens constructed of small framing members. As 

the cost of construction labor increased, the industry found that it was more economical to 

construct steel moment-frame buildings by moment-connecting a relatively small percentage 

of the beams and columns and by using larger members for these few moment-connected 

elements. The amount of strain demand placed on the connection elements of a steel moment 

frame is related to the span-to-depth ratio of the member. Therefore, as member sizes 

increased, strain demands on the welded connections also increased, making the connections 

more susceptible to brittle behavior. 

• In the 1960s and 1970s, when much of the initial research on steel moment-frame 

construction was performed, beams were commonly fabricated using A36 material. In the 

1980s, many steel mills adopted more modern production processes, including the use of 

scrap-based production. Steels produced by these more modern processes tended to include 

micro-alloying elements that increased the strength of the materials so that despite the 

common specification of A36 material for beams, many beams actually had yield strengths 

that approximated or exceeded that required for grade 50 material. As a result of this 

increase in base metal yield strength, the weld metal in the beam-flange-to-column-flange 

joints became under-matched, potentially contributing to its vulnerability. 

At this time, it is clear that in order to obtain reliable ductile behavior of steel moment-frame 

construction a number of changes to past practices in design, materials, fabrication, erection and 

quality assurance are necessary. The recommended criteria contained in this document, and the 

companion publications, are based on an extensive program of research into materials, welding 

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Chapter 1: Introduction Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

technology, inspection methods, frame system behavior, and laboratory and analytical 

investigations of different connection details. The recommended criteria presented herein are 

believed to be capable of addressing the vulnerabilities identified above and providing for frames 

capable of more reliable performance in response to earthquake ground shaking. 

1.4 Application 

This publication supersedes the design recommendations for new construction contained in 

FEMA-267, Interim Guidelines: Evaluation, Repair, Modification and Design of Welded Steel 

Moment Frame Structures, and the Interim Guidelines Advisories, FEMA-267A and FEMA267B. It is intended to be used as a basis for updating and revision of evaluation and 

rehabilitation guidelines and standards currently employed in steel moment-frame construction, 

in order to permit more reliable seismic performance in moment-resisting frame construction. 

This document has been prepared based on the provisions contained in FEMA-302 1997 NEHRP 

Recommended Provisions for Seismic Regulations for New Buildings and Other Structures 

(BSSC, 1997a), the 1997 AISC Seismic Specification (AISC, 1997), including supplements 

(AISC, 1999) and the 1998 AWS D1.1 Structural Welding Code - Steel, as it is anticipated that 

these documents form the basis for the current model building code, the 2000 edition of the 

International Building Code. Some users may wish to apply the recommendations contained 

herein to specific engineering projects, prior to the adoption of these recommendations by future 

codes and standards. Such users are cautioned to consider carefully any differences between the 

aforementioned documents and those actually enforced by the building department having 

jurisdiction for a specific project, and to adjust the recommendations contained in these 

guidelines accordingly. These users are also warned that these recommendations have not 

undergone a consensus adoption process. Users should thoroughly acquaint themselves with the 

technical data upon which these recommendations are based and exercise their own independent 

engineering judgment prior to implementing these recommendations. 

1.5 Overview 

The following is an overview of the general contents of chapters contained in these 

Recommended Criteria, and their intended use: 

• Chapter 2: General Requirements. This chapter, together with Chapter 3, is intended to 

indicate recommended supplements to the building code requirements for design of steel 

moment-frame buildings. These chapters include discussion of referenced codes and 

standards; design performance objectives; selection of structural systems; configuration of 

structural systems; and analysis of structural frames to obtain response parameters (forces 

and deflections) used in the code design procedures. Also included is discussion of an 

alternative, performance-based design approach that can be used at the engineer’s option, to 

design for superior or more reliable performance than is attained using the code basedapproach. Procedures for implementation of the performance-based approach are contained 

in Chapter 4. 

• Chapter 3: Connection Qualification. Steel moment frames can incorporate a number of 

different types of beam-column connections. Based on research conducted as part of this 

project, a number of connection details have been determined to be capable of providing 

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acceptable performance for use with different structural systems. These connections are 

termed prequalified. This chapter provides information on the limits of this prequalification 

for various types of connections and specific design and detailing recommendations for these 

prequalified connections. In some cases it may be appropriate to use connection details and 

designs which are different than the prequalified connections contained in this chapter, or to 

use one of the prequalified connection details outside the range of its prequalification. This 

chapter provides recommended criteria for project-specific qualification of a connection 

detail in such cases, as well as recommended procedures for new prequalifications for 

connections for general application. Reference to several proprietary connection types that 

may be utilized under license agreement with individual patent holders is also provided. 

When proprietary connections are used in a design, qualification data for such connections 

should be obtained directly from the licensor. 

• Chapter 4: Performance Evaluation. This chapter presents a simplified analytical 

performance evaluation methodology that may be used, at an engineer’s option, to 

determining the probable structural performance of regular, welded steel moment-frame 

structures, given the site seismicity. These procedures allow the calculation of a level of 

confidence that a structure will have less than a desired probability of exceeding either of two 

performance levels, an Immediate Occupancy level or a Collapse Prevention level. If the 

calculated level of confidence is lower than desired, a design can be modified and reevaluated for more acceptable performance, using these same procedures. 

• Appendix A: Detailed Procedures for Performance Evaluation. This appendix provides 

criteria for implementation of the detailed analytical performance evaluation procedures upon 

which the simplified procedures of Chapter 4 are based. Implementation of these procedures 

can permit more certain evaluation of the performance of a building to be determined than is 

possible using the simplified methods of Chapter 4. Engineers may find the application of 

these more detailed procedures beneficial in demonstrating that building performance is 

better than indicated by Chapter 4. Use of these procedures is required when a performance 

evaluation is to be performed for a building employing connections that have not been 

prequalified, or for a building that is irregular, as defined in FEMA-273. 

• References, Bibliography, and Acronyms. 

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2. GENERAL REQUIREMENTS 

2.1 Scope 

These Recommended Criteria apply to the seismic design of Special Moment Frames and 

Ordinary Moment Frames designed using the R, Cd, and W0 values given in Table 5.2.2, pages 

45-50, of FEMA-302. They do not apply to structures designed in accordance with the applicable 

Provisions of FEMA-302 for “Structural Steel Systems Not Specifically Detailed for Seismic 

Resistance”. These Recommended Criteria replace and supercede all design guidelines contained 

in FEMA-267, FEMA-267A, and FEMA-267B. 

This chapter presents overall criteria for the seismic design of steel moment frames for new 

buildings and structures. Included herein are general criteria on applicable references including 

codes, provisions and standards, recommended performance objectives, system selection, system 

analysis, frame design, connection design, specifications, quality control and quality assurance. 

2.2 Applicable Codes, Standards, and References 

Steel moment-frame systems should, as a minimum, be designed in accordance with the 

applicable provisions of the prevailing building code as supplemented by these Recommended 

Criteria. These Recommended Criteria are specifically written to be compatible with the 

requirements of FEMA-302 – NEHRP Recommended Provisions for Seismic Regulations for 

New Buildings and Other Structures. Where these Recommended Criteria are different from 

those of the prevailing code, it is intended that these Recommended Criteria should take 

precedence. The following are the major codes, standards and references referred to herein: 

FEMA-302 NEHRP Recommended Provisions for Seismic Regulations for New Buildings 

and Other Structures, 1997 Edition, Part 1 – Provisions (BSSC, 1997a) 

FEMA-303 NEHRP Recommended Provisions for Seismic Regulations for New Buildings 

and Other Structures, 1997 Edition, Part 2 – Commentary (BSSC, 1997b) 

AWS D1.1 Structural Welding Code, 1998 Edition (AWS, 1998) 

AISC Seismic Seismic Provisions for Structural Steel Buildings, April 15, 1997, (AISC, 1997) 

including Supplement No. 1, February 15, 1999 (AISC, 1999) 

AISC-LRFD Load and Resistance Factor Design Specifications for Structural Steel Buildings 

(AISC, 1993) 

AISC-Manual LRFD Manual of Steel Construction, Second Edition, 1998 (AISC, 1998b) 

FEMA-353 Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications (SAC, 2000d) 

FEMA-273 NEHRP Guidelines for the Seismic Rehabilitation of Buildings (ATC, 1997a) 

Commentary: The 1997 AISC Seismic Provisions (AISC, 1997) provide design 

requirements for steel moment-frame structures. FEMA-302 adopts the AISC 

Seismic Provisions by reference as the design provisions for seismic-force2-1
Recommended Seismic Design 

Chapter 2: General Requirements Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

resisting systems of structural steel. The International Building Code is based 

generally on the FEMA-302 Provisions, and incorporates design requirements for 

steel structures primarily based on the AISC Provisions. These Recommended 

Criteria are written to be compatible with the 1997 AISC Seismic Provisions and 

FEMA-302 Provisions and reference is made to sections of those documents 

where appropriate herein. 

2.3 Basic Design Approach 

The recommended design approach consists of the following basic steps: 

Step 1: Select a structural system type and frame configuration in accordance with Section 2.5 of 

these Recommended Criteria. 

Step 2: Select preliminary frame member sizes and perform a structural analysis for earthquake 

loading and frame adequacy using the applicable R, Cd and W0 values, strength criteria, 

drift limits, and redundancy requirements of FEMA-302, as supplemented by Section 2.9 

of these Recommended Criteria. 

Step 3: Select an appropriate connection type, in accordance with Section 2.5.3 of these 

Recommended Criteria. Connections may be prequalified, project qualified, or 

proprietary, as indicated in Chapter 3 of these Recommended Criteria. 

Step 4: Perform an analysis in accordance with Sections 2.7 and 2.8 of these Recommended 

Criteria, considering the effects (if any) of the selected connection type on frame 

stiffness and behavior, to confirm the adequacy of member sizing to meet the applicable 

strength, drift, and stability limitations. 

Step 5: Confirm or revise the member sizing based on the connection type selected and 

following Sections 2.9 and 3.2 of these Recommended Criteria. Return to Step 4, if 

necessary. 

Step 6: Complete the design of the connections, in accordance with Chapter 3 of these 

Recommended Criteria. 

As an option, when it is desired to design for specific performance, rather than simply 

achieving code compliance, a Performance Evaluation following the guidelines of Chapter 4 may 

be performed. 

Commentary: This section outlines the basic steps recommended for design 

intended to meet the minimum criteria of the building code. Since the 1994 

Northridge earthquake, the 1997 AISC Seismic Provisions have required that 

laboratory test data be submitted to demonstrate that connection detailing will be 

capable of adequate service. With the publication of these Recommended 

Criteria, and the establishment of a series of prequalified connection details, it is 

intended that substantiation of connection detailing by reference to laboratory 

test data will not be required for most design applications. However, design 

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Criteria for New Steel FEMA-350 

procedures for some types of prequalified connections entail significant 

calculation. 

The optional Performance Evaluation procedures contained in Chapter 4 and 

Appendix A of these Recommended Criteria need not be applied to designs 

intended only to meet the requirements of the building code. Regular, wellconfigured Special Moment Frame and Ordinary Moment Frame structures 

designed and constructed in accordance with FEMA-302, and building code 

requirements as supplemented by these Recommended Criteria, are expected to 

provide a high level of confidence of being able to resist collapse under Maximum 

Considered Earthquake demands. Section 2.4 of these Recommended Criteria 

and FEMA-303 provide additional information on this performance goal. 

Structures with significant irregularity, low levels of redundancy, or poor 

configuration may not be capable of such performance. The Performance 

Evaluation procedures of Chapter 4 and Appendix A may be used to confirm the 

capability of such structures to meet the performance intended by the building 

code, or may be used to implement performance-based designs intended to meet 

higher performance objectives. 

2.4 Design Performance Objectives 

Under FEMA-302, each building and structure must be assigned to one of three Seismic Use 

Groups (SUGs). Buildings are assigned to the SUGs based on their intended occupancy and use. 

Most commercial, residential and industrial structures are assigned to SUG I. Buildings occupied 

by large numbers of persons or by persons with limited mobility, or that house large quantities of 

potentially hazardous materials are assigned to SUG II. Buildings that are essential to 

postearthquake disaster response and recovery operations are assigned to SUG III. Buildings in 

each of SUG II and III are intended to provide better performance, as a group, than buildings in 

SUG I. As indicated in FEMA-303, buildings designed in accordance with the provisions for 

each SUG are intended, as a minimum, to be capable of providing the performance indicated in 

Figure 2-1. 

The FEMA-302 provision attempts to obtain these various performance characteristics 

through regulation of system selection, detailing requirements, design force levels, and 

permissible drift. This regulation is based on the SUG, the seismicity of the region containing 

the building site, and the effect of site-specific geologic conditions. All structures should, as a 

minimum, be assigned to an appropriate SUG, in accordance with the building code, and be 

designed in accordance with the applicable requirements for that SUG. 

Although the FEMA-303 Commentary to FEMA-302 implies that buildings designed in 

accordance with the requirements for the various SUGs should be capable of providing the 

performance capabilities indicated in Figure 2-1, FEMA-302 does not contain direct methods to 

evaluate and verify the actual performance capability of structures, nor does it provide a direct 

means to design for performance characteristics other than those implied in Figure 2-1, should it 

be desired to do so. It is believed, based on observation of the performance of modern, code2-3
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Chapter 2: General Requirements Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

conforming construction in recent earthquakes, that FEMA-302 provides reasonable reliability 

with regard to attaining Life Safe performance for SUG-I structures subjected to design events, as 

indicated in Figure 2-1. However, the reliability of FEMA-302 with regard to the attainment of 

other performance objectives for SUG-I structures, or for reliably attaining any of the 

performance objectives for the other SUGs seems less certain and has never been quantified or 

verified. 

Building Performance Levels 

Operational 

Immediate 

Occupancy 

Life 

Safe 

Near 

Collapse 

Ground Motion Levels 

Maximum 

Considered 

Earthquake 

(2% - 50 years) 

(50% - 50 years) Performance for Group I Buildings

Performance for Group III Buildings

Performance for Group II Buildings

Frequent 

Earthquakes 

Design 

Earthquake 

(2/3 of MCE) 

Figure 2-1 NEHRP Seismic Use Groups (SUG) and Performance 

Chapters 2 and 3 of these Recommended Criteria present code-based design 

recommendations for steel moment-frame buildings. All buildings should, as a minimum, be 

designed in accordance with these recommendations. For buildings in which it is desired to 

attain other performance than implied by the code, or for which it is desired to have greater 

confidence that the building will actually be capable of attaining the desired performance, the 

procedures in Chapter 4 and Appendix A may be applied. 

Commentary: FEMA-302 includes three types of steel moment frames , two of 

which are incorporated in these Recommended Criteria. The three types are: 

Special Moment Frames (SMF), Intermediate Moment Frames (IMF), and 

Ordinary Moment Frames (OMF). Building code provisions for SMF systems 

strictly regulate building configuration, proportioning of members and 

connection detailing in order to produce structures with superior inelastic 

response capability. Provisions for OMF systems have less control on these 

design features and therefore, as a class, OMF structures are expected to have 

poorer inelastic response capability than SMF systems. Following the 1994 

Northridge earthquake, the building code was amended to include substantial 

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additional requirements for SMF system design and construction, resulting in an 

increase in the development cost for such structures. In 1997, the IMF system 

was added to FEMA-302 and the AISC Seismic Provisions to provide an 

economical alternative to SMF construction for regions of moderate seismicity. 

Studies conducted under this project have indicated that the inelastic response 

demands on IMF systems are similar to those for SMF systems and that, 

therefore, the reduction in design criteria associated with the IMF system was not 

justified. Consequently, only Special Moment Frame and Ordinary Moment 

Frame systems are included herein. These systems are described in more detail 

in Section 2.5. In FEMA-302, a unique R value and Cd factor are assigned to 

each of these systems, as are height limitations and other restrictions on use. 

Regardless of the system selected, FEMA-302 implies that structures designed to 

meet the requirements therein will be capable of meeting the Collapse Prevention 

performance level for a Maximum Considered Earthquake (MCE) ground motion 

level and will provide Life Safe performance for the Design Basis Earthquake 

(DBE) ground motion that has a severity 2/3 that of MCE ground motion. This 

2/3 factor is based on an assumption that the Life Safety performance on which 

earlier editions of the NEHRP Recommended Provisions were based inherently 

provided a minimum margin of 1.5 against collapse. Except for sites located 

within a few kilometers of known active faults, the MCE ground motion is 

represented by ground shaking response spectra that have a 2% probability of 

exceedance in 50 years (approximately 2500-year mean return period). For sites 

that are close to known active faults, the MCE ground motion is taken either as 

this 2%/50-year spectrum, or as a spectrum that is 150% of that determined from 

a median estimate of the ground motion resulting from a characteristic event on 

the nearby active fault, whichever is less. 

The FEMA-302 Provisions define classes of structures for which performance 

superior to that described above is mandated. Additionally, individual building 

owners may desire a higher level of performance. The FEMA-302 Provisions 

attempt to improve performance for SUG-II and SUG-III structures, (1) through 

use of an occupancy importance factor that increases design force levels, and 

therefore reduces the amount of ductility a structure must exhibit to withstand 

strong ground shaking, and (2) through specification of more restrictive drift 

limits than those applied to SUG-I structures. This combination of increased 

design forces and more restrictive drift limitations leads to substantially greater 

strength in systems such as SMFs, the design of which is governed by drift. 

The FEMA-302 R factors, drift limits, and height limitations, as well as the 

inelastic rotation capability requirements corresponding to the R value for each 

moment-frame type (SMF, IMF, or OMF), are based more on historical precedent 

and judgment than they are on analytical demonstration of adequacy. In the 

research program on which these Criteria are based, an extensive series of 

nonlinear analytical investigations has been conducted to determine the drift 

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Chapter 2: General Requirements Moment-Frame Buildings 

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demands on structures designed in accordance with the current code when 

subjected to different ground motions, and for a variety of assumed hysteretic 

behaviors for connections. The results of these investigations have led to the 

conclusion that some of the FEMA-302 Provisions and 1997 AISC Seismic 

Provisions were not capable of reliably providing the intended performance. 

These Recommended Criteria directly modify those Provisions so as to increase 

the expected reliability of performance to an acceptable level. On the basis of 

these analytical studies, it is believed that regular, well-configured structures 

designed in accordance with these Recommended Criteria and constructed in 

accordance with FEMA-353, provide in excess of 90% confidence of being able to 

withstand Maximum Considered Earthquake demands without global collapse 

and provide mean confidence of being able to withstand such ground motion 

without local structural failure. 

It should be recognized that application of the modifications suggested in 

these Recommended Criteria, while considered necessary to provide this level of 

confidence with regard to achieving the indicated performance for momentresisting frames, may result in such systems having superior performance 

capabilities relative to some other systems, the design provisions for which do not 

have a comparable analytical basis. In other words, the design provisions 

contained in FEMA-302 for some other structural systems, both of steel and of 

other construction materials, may inherently provide a lower level of assurance 

that the resulting structures will be able to provide the intended performance. 

The three classes of steel moment-frame systems contained in FEMA-302 are 

themselves not capable of providing uniform performance. OMFs will typically 

be stronger than either IMFs or SMFs, but can have much poorer inelastic 

response characteristics. The result of this is that OMFs should be able to resist 

the onset of damage at somewhat stronger levels of ground shaking than is the 

case for either IMFs or SMFs. However, as ground motion intensity increases 

beyond the damage threshold for each of these structural types, it would be 

anticipated that OMFs would present a much greater risk of collapse than would 

IMFs, which in turn, would present a more significant risk of collapse then SMFs. 

For these reasons, FEMA-302 places limitations on the applicability of these 

various structural systems depending on a structure’s height and the seismic 

hazard at the site. 

Refer to Chapter 4 for more detailed discussion of recommended performance 

objectives and their implications. 

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2.5 System Selection 

2.5.1 Configuration and Load Path 

Every structure should be provided with a complete lateral and vertical seismic-forceresisting system, capable of transmitting inertial forces from the locations of mass throughout the 

structure to the foundations. For steel moment-frame structures, the load path includes the floor 

and roof diaphragms, the moment-resisting frames, the foundations, and the various collector 

elements that interconnect these system components. 

To the extent possible, the structural system should have a regular configuration without 

significant discontinuities in stiffness or strength and with the rigidity of the structural system 

distributed uniformly around the center of mass. 

Commentary: The importance of maintaining regularity in structural systems can 

not be overemphasized. The analytical investigations of structural performance 

conducted as part of this project were limited to regular structural systems. 

Irregularities in structural systems can result in concentration of deformation 

demands on localized portions of a structure, and early development of P-D 

instabilities. FEMA-302 includes significant limitations on structural 

irregularity, particularly for structures in Seismic Design Categories D, E and F. 

However, it was not possible, within the scope of this project, to determine if these 

limitations are sufficient to ensure that the intended performance capability is 

achieved and this should be the subject of future investigations. 

Structures categorized as regular under FEMA-302 may not actually behave 

in a regular manner. FEMA-302 categorizes a multistory buildings as being 

regular if the vertical distribution of lateral stiffness and strength is uniform. 

Thus, a structure with equal lateral stiffness and strength in every story would be 

categorized as regular. However, such structures would not actually behave as 

regular structures when responding to strong ground motion. Instead such 

structures would develop large concentrations of inelastic behavior and 

deformation at the lower stories of the structure. To provide true strength and 

stiffness regularity in multistory structures, it is necessary to maintain uniform 

ratios of (1), lateral strength to tributary mass, and (2), lateral stiffness to 

tributary mass, for each story of the structure, where tributary mass may be 

considered as that portion of the structure’s mass supported at and above the 

story. 

2.5.2 Structural System Selection 

The moment frame may be designed either as an SMF or OMF. The selection of momentframe type should be governed by the prevailing code and by the project conditions. 

Consideration should be given to using Special Moment Frames whenever conditions permit. 

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Chapter 2: General Requirements Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

Commentary: FEMA-302 defines three types of steel moment frames: Special 

Moment Frames (SMF), Intermediate Moment Frames (IMF), and Ordinary 

Moment Frames (OMF). Detailing and configuration requirements are specified 

for each of these three systems to provide for different levels of ductility and 

global inelastic response capability, varying from highest in SMFs to lowest in 

OMFs. IMF systems have intentionally been omitted from these Recommended 

Criteria because nonlinear analyses of buildings designed to the criteria for IMF 

systems contained in FEMA-302 have indicated that the inelastic demands for 

these structures are nearly as large as those for SMF structures. Therefore, it is 

not possible to justify on technical grounds the use of the relaxed detailing 

criteria provided for IMFs in FEMA-302 unless more restrictive design force 

levels and drift criteria are also specified in order to limit the amount of inelastic 

demand these structures may experience. Rather than developing such criteria, it 

was decided to omit this system, which had only recently been introduced into the 

building codes, from further consideration. 

Ordinary Moment Frames are relatively strong (compared to SMFs) but have 

much less ductility. As a result, Ordinary Moment Frame structures, as a class, 

would be anticipated to have less damage than SMFs for moderate levels of 

ground shaking and significantly more damage than SMFs for severe levels of 

ground shaking. In recognition of this, FEMA-302 places limitations on the 

height, occupancy and ground motion severity for which Ordinary Moment Frame 

systems can be used. In recognition of the superior performance characteristics 

of SMF systems when subjected to high-intensity ground shaking, it is 

recommended that designers consider their use, even when IMF or OMF systems 

are permitted under the building code. 

2.5.3 Connection Type 

Moment-resisting connections in SMFs and OMFs, except connections in OMFs designed to 

remain elastic under design level earthquake ground shaking, should be demonstrated by test and 

by analysis to be capable of providing the minimum levels of interstory drift angle capacity 

specified in Section 3.9 of these Recommended Criteria. Interstory drift angle is that portion of 

the interstory drift ratio in a frame resulting from flexural deformation of the frame elements, as 

opposed to axial deformation of the columns, as indicated in Figure 2-2. Sections 3.5, 3.6 and 

3.7 present details and design procedures for a series of connections that are recommended as 

prequalified to meet the criteria of Section 3.9 without further analysis or testing, when used 

within the indicated limits applicable to each connection type. 

Commentary: FEMA-302 and the 1997 AISC Seismic Provisions set minimum 

strength criteria for connections. In addition, except for connections in OMFs 

that are designed to remain elastic, the 1997 AISC Seismic Provisions require 

that connections be demonstrated capable of providing minimum levels of 

rotational capacity. The 1997 AISC Seismic Provisions uses plastic rotation angle 

as the performance parameter by which connections are qualified. In these 

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Criteria for New Steel FEMA-350 

Recommended Criteria, interstory drift angle is used instead. This is because this 

parameter, (1) seems to be stable with regard to prediction of frame performance, 

(2) is closely related to plastic rotation angle, (3) is less ambiguous with regard to 

definition, and (4) is a quantity that is easily determined from the results of 

standard frame analyses using either linear or nonlinear methods. 

Drift angle 

Deformed Deformed

shape

Undeformed Undeformed

shape

Figure 2-2 Interstory Drift Angle 

Figure 2-2 illustrates the interstory drift angle, for a frame with fully 

restrained (FR) connections and rigid panel zones. Prior to lateral deformation, 

the beam and column are joined at right angles to each other. Under elastic 

deformation, the column and beam will remain joined at right angles and the 

beam will rotate in double curvature between the two columns. The interstory 

drift angle is measured as the angle between the undeformed vertical axis of the 

column and the deformed axis of the column at the center of the beam-column 

joint. For the idealized FR frame with rigid panel zones, shown in the figure, this 

same angle will exist between the undeformed horizontal axis of the beam and the 

deformed axis of the beam, at the beam-column connection. In FEMA-273, this 

angle is termed the chord angle and is used as the parameter for determining 

beam-column connection performance. However, for frames with panel zones 

that are not rigid, frames with partially restrained connections, or frames that 

exhibit plasticity at the connection, the chord angle of the beam will not be 

identical to the interstory drift angle. For such frames, the interstory drift angle, 

reduced for the effects of axial column elongation, is a better measure of the total 

imposed rotation on all elements of the connection, including panel zones and 

connection elements, and is used as the basis of these Recommended Criteria. 

2.5.4 Redundancy 

Structures assigned to Seismic Design Categories D, E, and F of FEMA-302 shall be 

provided with sufficient bays of moment-resisting framing to satisfy the redundancy 

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Chapter 2: General Requirements Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

requirements of those Provisions. In addition, the strength of members of the seismic-forceresisting system shall be evaluated for adequacy to resist horizontal earthquake forces that are 

factored by the redundancy factor r in accordance with the load combinations of FEMA-302. 

Commentary: There are several reasons why structures with some redundancy in 

their structural systems should perform better than structures without such 

redundancy. The basic philosophy underlying the design provisions of FEMA302 is to permit substantial inelastic behavior in frames under ground shaking of 

the severity of the design earthquake or more severe events. Under such 

conditions, occasional failures of elements may occur. Structures that have 

nonredundant seismic-force-resisting systems could potentially develop instability 

in the event of failure of one or more elements of the system. Redundant 

structures, on the other hand, would still retain some significant amount of lateral 

resistance following failure of a few elements. 

Another significant advantage of providing redundant framing systems is that 

the use of a larger number of frames to resist lateral forces often permits the size 

of the framing elements to be reduced. Laboratory research has shown that 

connection ductile capacity generally increases as the size of the framing 

elements decreases. 

FEMA-302 includes a redundancy factor r with values between 1.0 and 1.5, 

which is applied as a load factor on calculated earthquake forces for structures 

categorized as Seismic Design Category D, E, or F. Less redundant systems 

(frames with fewer participating beams and columns) are assigned higher values 

of the redundancy factor and therefore must be designed to resist higher design 

forces to compensate for their lack of redundancy. Minimum permissible levels of 

redundancy are set, through lower-bound values specified for the redundancy 

factor, for structures located in regions of high seismic risk. 

The maximum permitted r values given in FEMA-302 were based only on the 

judgment of the writers of that document. They should not be construed as ideal 

or optimum values. Designers are encouraged to incorporate as much 

redundancy as is practical into steel moment-frame buildings. 

2.5.5 Frame Beam Spans 

The connection prequalification data provided for each prequalified connection in Chapter 3 

includes specification of the minimum beam-span-to-depth ratio for which the connection is 

prequalified. Span-to-depth ratios for beams in moment frames should equal or exceed the 

minimum span-to-depth ratio applicable to the connection type being used, unless projectspecific qualification testing is performed as described in Section 3.9, or other rational analysis is 

employed to demonstrate that hinge rotations or bending strains will not exceed those for which 

the connection is prequalified. 

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Where the effective span for a frame beam (distance between points of plastic hinging of the 

beam) is such that shear yielding of the beam will occur before flexural yielding, the web of the 

beam shall be detailed and braced as required by the 1997 AISC Seismic Provisions for long links 

in eccentric braced frames. 

Commentary: In determining the layout of moment frames, it should be 

recognized that excessively short spans can affect both frame and connection 

behavior. Possible effects include the following: 

1. For connection types that move the hinge significantly away from the column face, 

the plastic rotation demand at the hinge will be significantly larger than the frame 

interstory drift angle, due to geometric effects. 

2. The steeper moment gradient resulting from the shorter spans will decrease the 

length of the beam hinge, requiring that the beam develop greater bending strains to 

accommodate the same interstory drift angle. 

3. If the effective span length becomes too short, shear yielding of the beam, rather than 

flexural yielding, will control inelastic behavior. 

Most testing of prequalified connections performed under this project used 

configurations with beam spans of about 25 feet. Most tested beams were either 

W30 or W36, so that span-to-depth ratios were typically in the range of 8 to 10. 

Refer to FEMA-355D, State of the Art Report on Connection Performance for 

more information on the effects of short spans. 

2.6 Structural Materials 

2.6.1 Material Specifications 

Structural steel should conform to the specifications and grades permitted by the 1997 AISC 

Seismic Provisions, as modified by FEMA-353, and as indicated in the specific connection 

prequalifications, unless a project-specific qualification testing program is performed to 

demonstrate acceptable performance of alternative materials. 

Commentary: Under the 1997 AISC Seismic Provisions, rolled shapes used in 

OMF or SMF applications may conform to the ASTM A36, A572 or A913 

specifications. In the 1980s, it was common practice in some regions to design 

moment frames with columns conforming to the ASTM A572 Grade 50 

specification and with beams conforming to the ASTM A36 specification, in order 

to obtain frames economically with strong columns and weak beams. During the 

1990s, however, the steel production industry in the United States has undergone 

a significant evolution, with many of the older mills being replaced by newer mills 

that use scrap-based production processes. These newer mills routinely produce 

higher strength steel than did the older mills. Since the A36 and A572 

specifications do not place an upper bound on material strength, much of the steel 

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Chapter 2: General Requirements Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

shipped by these mills, particularly for material ordered as conforming to the A36 

specification, is much stronger than the minimum strength controlled by the 

specification, and use of the combination of A36 and A572 materials to provide 

for strong-column-weak-beam conditions will not reliably achieve this goal. In 

1997, ASTM introduced a new A992 specification to address this problem. The 

A992 specification is similar to the ASTM A572, Grade 50 specification, except 

that maximum as well as minimum yield strengths are specified to provide for 

more controlled design conditions. In addition, the A992 specification includes 

increased control on trace elements and can be more weldable than some A572 

steels. It is recommended that either A992 or A913 steel be used in SMF 

applications. 

2.6.2 Material Strength Properties 

The strength of materials shall be taken as indicated in the AISC Seismic Provisions and as 

modified by these Recommended Criteria. Where these Recommended Criteria require the use 

of “expected strength,” this shall be the quantity Ry Fy as indicated in the AISC Seismic 

Provisions. The value of Ry for material conforming to ASTM A992 shall be the same as for 

material conforming to ASTM A572 Grade 50. Where these Recommended Criteria require the 

use of lower-bound strength, or specified strength, the minimum specified value of the yield 

strength Fy as indicated in the applicable ASTM specification shall be used. 

Commentary: The AISC Seismic Provisions specify values of Ry for various 

materials as indicated in Table 2-1. The quantity RyFy is intended to approximate 

the mean value of the yield strength of material produced to a given specification 

and grade. The AISC Seismic Provisions permit other values of Ry to be used, if 

the value of the expected mean yield strength Fye is determined by appropriate 

testing. 

Table 2-1 Values of Ry for Various Material Grades 

Material Specification Ry 

ASTM A36 1.5 

ASTM A572 Gr. 42 1.3 

Other Specifications 1.1 

As part of the program of investigations conducted in support of the 

development of these Recommended Criteria, studies of the statistical variation in 

strength properties of rolled sections of Grade 50 steel were conducted. These 

studies indicate that the 1.1 value for Ry is a good representation of the mean 

value of yield strength when applied to the webs of cross sections. The flexural 

properties of structural steel, however, are more closely related to the yield 

strength of the flanges of rolled shapes, which tend to have somewhat lower 

strength than do the webs. When applied to calculations of the flexural strength 

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of beams, the use of an Ry value of 1.1 actually approximates a mean-plus-onestandard-deviation value. Since values of expected strength are used to estimate 

the amount of force that can be delivered to adjacent connected elements, the use 

of this conservative value is appropriate. More information on the statistical 

variation of steel strength may be found in FEMA-355A, State of the Art Report 

on Base Metals and Fracture. 

2.7 Structural Analysis 

An analysis should be performed for each structure to determine the distribution of forces and 

deformations under code-specified ground motion and loading criteria. The analysis should 

conform, as a minimum, to the code-specified criteria for the equivalent lateral force method or 

the modal response spectrum method, as applicable. 

Chapter 4 provides guidance on analysis methods that can be used as part of the Performance 

Evaluation approach for steel moment-frame structures. 

Commentary: Seismic design forces for low-rise and mid-rise buildings without 

major irregularities have traditionally been determined by using the simple 

“equivalent lateral force” method prescribed by the codes. Such methods are 

incorporated in FEMA-302 and are permitted to be used for structures designated 

as regular, and up to 240 feet in height. Buildings that are over 5 stories or 65 

feet in height and have certain vertical irregularities, and all buildings over 240 

feet in height, require use of dynamic (modal or response history) analysis. The 

use of inelastic response history or nonlinear static analysis is also permitted by 

some codes though few guidelines are provided in codes on how to perform or 

apply such an analysis. Projects incorporating nonlinear response-history 

analysis should be conducted in accordance with the Performance Evaluation 

provisions of Chapter 4. For such applications, structures should be 

demonstrated as capable, with 90% confidence, of providing Collapse Prevention 

performance for MCE hazards based on considerations of global behavior and 

column adequacy. A 50% confidence level should be demonstrated for connection 

behavior. 

2.8 Mathematical Modeling 

2.8.1 Basic Assumptions 

In general, a steel moment-frame building should be modeled, analyzed and designed as a 

three-dimensional assembly of elements and components. Although two-dimensional models 

may provide adequate design information for regular, symmetric structures and structures with 

flexible diaphragms, three-dimensional mathematical models should be used for analysis and 

design of buildings with plan irregularity as defined by FEMA-302. The two-dimensional 

modeling, analysis, and design of buildings with stiff or rigid diaphragms is acceptable, if 

torsional effects are either sufficiently small to be ignored, or are captured indirectly. 

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Vertical lines of framing in buildings with flexible diaphragms may be individually modeled, 

analyzed and designed as two-dimensional assemblies of components and elements, or a threedimensional model may be used, with the diaphragms modeled as flexible elements. 

Explicit modeling of connections is required only for nonlinear procedures and only if (1) the 

connection is weaker than the connected components, or (2) the flexibility of the connection 

results in a significant increase in the relative deformation between connected components. 

Additional guidance in using these methods is found in Chapter 4. 

Commentary: A finite-element model will provide information on forces and 

deformations only at places in the structure where a modeling element is inserted. 

When nonlinear deformations are expected in a structure, the designer must 

anticipate the location of the plastic hinges and insert nonlinear finite elements at 

these locations if the inelastic behavior is to be captured by the model. Additional 

information is found in Chapter 4. 

2.8.2 Model Configuration 

The analytical model should accurately account for the stiffness of frame elements and 

connections and other structural and nonstructural elements that may affect this stiffness. This 

section presents basic recommendations for analyses intended to meet the requirements of 

FEMA-302. More detailed modeling guidelines for the purposes of performance evaluation are 

presented in Chapter 4. Chapter 3 presents specific modeling guidelines for various prequalified 

connections, referred to by the guidelines of Section 2.8, and Chapter 4. 

2.8.2.1 Regularity 

Classification of a building as irregular, and analysis limitations based on regularity are 

discussed in FEMA-302. Such classification should be based on the plan and vertical 

configuration of the framing system, using a mathematical model that considers relevant 

structural members. 

2.8.2.2 Elements Modeled 

For the purpose of determining the adequacy of the structure to meet the strength and drift 

requirements of FEMA-302, only participating elements of the seismic-force-resisting system 

shall be included in the analytical model. When nonstructural or nonparticipating elements of the 

seismic-force-resisting system have significant influence on the stiffness or distribution of 

seismic forces within the elements of the seismic-force-resisting system, separate analyses should 

be performed to evaluate the effect of these elements on (1) the distribution of deformations and 

member forces, and (2) overall building performance. 

Commentary: In order to comply with the requirements of FEMA-302, it is 

necessary that the seismic-force-resisting system be capable of resisting the 

design seismic forces without participation of other elements. However, steel 

moment-frame structures are inherently flexible. Rigid supported elements 

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including architectural wall systems, ramped floors, and large mechanical 

equipment items can affect both the stiffness of the structure and the distribution 

of forces within the structure. The best practice in the design and detailing of 

steel moment-frame structures is to detail elements that are not part of the 

seismic-force-resisting system such that they are isolated from participating in the 

resistance of earthquake-induced frame drifts. For those cases when such 

isolation is not possible, the effect of these elements on the behavior of the frame 

should be considered in the design. 

FEMA-302 does not permit consideration of elements that are not part of the 

primary lateral-force-resisting system as effective in meeting the strength and 

stiffness requirements of the provisions. However, in many steel moment-frame 

structures, framing provided only to resist gravity loads can provide substantial 

additional stiffness and strength. It is recommended that the effect of these 

nonparticipating structural elements be considered when performing analyses in 

support of performance evaluations, conducted in accordance with Chapter 4 of 

these Recommended Criteria. 

2.8.2.3 Connection Stiffness 

For frames with fully restrained connections, it shall be permissible to model the frame using 

centerline-to-centerline dimensions for the purpose of calculating stiffnesses of beams and 

columns. Alternatively, when justified by appropriate analytical or test data, more realistic 

assumptions that account for the stiffness of panel zones and connections may be used. In either 

case, calculation of beam moments and shears should be performed at the face of the column. 

For linear analysis of structures with partially restrained connections, beams should be 

modeled with an equivalent EI, using the method shown in Chapter 5 of FEMA-273. Chapter 3 

of these Recommended Criteria provides guidelines for estimating connection stiffness 

parameters for use in this procedure for the various prequalified partially restrained connections. 

For nonlinear analysis of frames with partially restrained connections, the nonlinear forcedeformation characteristics of the connections should be directly modeled. 

Commentary: In analytical studies of moment-resisting frame behavior (FEMA355C) conducted in support of the development of these Recommended Criteria, it 

has been demonstrated that panel-zone deformations have little effect on 

analytical estimates of drift and need not be explicitly modeled, provided the 

panel zones are not excessively weak. Inelastic analyses of frames designed in 

accordance with these Recommended Criteria indicate that explicit modeling of 

panel zone shear strength and flexibility results in similar, albeit slightly smaller 

estimates of interstory drift than is obtained from models in which panel zones are 

not modeled and center-line-to-center-line framing dimensions are used. 

Therefore, this document recommends use of the simpler approach, in which 

panel zones are neglected in the model and center-line-to-center-line framing 

dimensions are used. It is permissible to use realistic assumptions for the 

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Chapter 2: General Requirements Moment-Frame Buildings 

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stiffness of panel zones, to modify the effective flexural span length of beams and 

columns, provided that such assumptions are based on appropriate data. Some 

connections, such as large haunches or slide plates, may significantly increase 

frame stiffness, meriting the inclusion of their effects in the analytical model. 

Additional discussion on modeling considerations, including methods to model 

connections and panel zones explicitly may be found in FEMA-355C, State of the 

Art Report on Systems Performance. 

2.8.3 Horizontal Torsion 

The effects of horizontal torsion must be considered, as in FEMA-302. The total torsional 

moment at a given floor level includes the following two torsional moments: 

a. Actual torsion: the moment resulting from the eccentricity between (1) the centers of mass at 

all floors above and including the given floor, and (2) the center of rigidity of the vertical 

seismic elements in the story below the given floor, and 

b. Accidental torsion: an accidental torsional moment produced by an artificial horizontal offset 

in the centers of mass, at all floors above and including the given floor, equal to a minimum 

of 5% of the horizontal dimension at the given floor level measured perpendicular to the 

direction of the applied load. 

When the effects of torsion are investigated, the increased forces and displacements from 

horizontal torsion should be evaluated and considered for design. The effects of torsion cannot 

be used to reduce force and deformation demands on components and elements. 

Commentary: Actual torsion that is not apparent in an evaluation of the center of 

rigidity and center of mass in an elastic stiffness evaluation can develop during 

nonlinear response of the structure if yielding develops in an unsymmetrical 

manner. For example, if the frames on the east and west sides of a structure have 

similar elastic stiffness the structure may not have significant torsion during 

elastic response. However, if the frames on the east side of the structure yield 

significantly sooner than the framing on the west side, then inelastic torsion will 

develop. Although the development of such inelastic torsion can be a serious 

problem, FEMA-302 does not address these phenomena. Designers can reduce 

the potential for severe inelastic torsion by providing framing layouts that have 

both stiffness and strength as symmetrical as possible about the center of mass. 

2.8.4 Foundation Modeling 

Foundations should generally be modeled as unyielding. Soil-structure interaction may be 

modeled as permitted by the building code. Assumptions for the extent of fixity against rotation 

provided at the base of columns should realistically account for the relative rigidities of the frame 

and foundation system, including soil compliance effects, and the detailing of the column base 

connections. 

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Commentary: Most steel moment frames can be adequately modeled by assuming 

that the foundation provides rigid support for vertical loads. However, the 

flexibility of foundation systems (and the attachment of columns to those systems) 

can significantly alter the flexural stiffness at the base of the frame. Where 

relevant, these factors should be considered in developing the analytical model. 

2.8.5 Diaphragms 

Floor and roof diaphragms transfer earthquake-induced inertial forces to vertical elements of 

the seismic framing system. Connections between diaphragms and vertical seismic framing 

elements must have sufficient strength to transfer the maximum calculated diaphragm shear 

forces to the vertical framing elements. Requirements for design and detailing of diaphragm 

components are given in FEMA-302. 

Diaphragms should be classified as flexible, stiff, or rigid in accordance with FEMA-302. 

For buildings with steel moment-frame systems, most floor slabs with concrete fill over metal 

deck may be considered to be rigid diaphragms. Floors or roofs with plywood diaphragms 

should be considered flexible. The flexibility of unfilled metal deck, and concrete slab 

diaphragms with large openings should be considered in the analytical model. 

Mathematical models of buildings with diaphragms that are not rigid should be developed 

considering the effects of diaphragm flexibility. For buildings with flexible diaphragms at each 

floor level, the vertical lines of seismic framing may be designed independently, with seismic 

masses assigned on the basis of tributary area. 

2.8.6 P-D Effects 

The structure shall be investigated to ensure that lateral drifts induced by earthquake response 

do not result in a condition of instability under gravity loads. At each story, the quantity Yi 

should be calculated for each direction of response, as follows: 

PRd Yi = i i (2-1) V hyii 

where: 

Pi = portion of the total weight of the structure including dead, permanent live, and 

25% of transient live loads acting on all of the columns within story level i, kips, 

R = response modification coefficient obtained applicable to the structural system and 

used to determine the design seismic forces 

di = calculated lateral drift at the center of rigidity of story i, when the design seismic 

forces are applied in the direction under consideration, inches, 

Vyi = total plastic lateral shear force in the direction under consideration at story i, 

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hi = height of story i, which may be taken as (1) the distance between the centerline of 

floor framing at each of the levels above and below, (2) the distance between the 

top of floor slabs at each of the levels above and below, or (3) the distance 

between similar common points of reference. 

Commentary: The quantity Yi is the ratio of the effective story shear produced by 

first order P-D effects at the calculated story drift to the maximum restoring force 

in the structure. When this ratio has a value greater than 1.0, the structure does 

not have enough strength to resist the P-D induced shear forces and unless 

restrained, will collapse in a sidesway mechanism. If the ratio is less than 1, the 

restoring force in the structure exceeds the story shear due to P-D effects and 

unless additional displacement is induced or lateral forces applied, the structure 

should not collapse. 

The plastic story shear quantity, Vyi, should be determined by methods of 

plastic analysis. In a story in which all beam-column connections meet the 

strong-column-weak-beam criterion, the same number of moment-resisting bays is 

present at the top and bottom of the frame and the strength of moment-connnected 

girders at the top and bottom of the frame is similar, Vyi may be approximately 

calculated from the equation: 

n 

2� M pG j 

Vyi = 

j =1 (2-2) hi 

where: 

MpGj = the plastic moment capacity of each girder “j” participating in the 

moment-resisting framing at the floor level on top of the story, and 

n = the number of moment-resisting girders in the framing at the floor 

level on top of the story. 

In any story in which all columns do not meet the strong-column-weak-beam 

criterion, the plastic story shear quantity, Vyi may be calculated from the 

equation: 

m 

2� M pCk 

Vyi = k =1 

h (2-3) 

i 

where: 

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m = the number of columns in moment-resisting framing in the story 

under consideration, and 

MpCk = the plastic moment capacity of each column “k”, participating in 

the moment-resisting framing, considering the axial load present 

on the column. 

For other conditions, the quantity Vyi must be calculated by plastic mechanism 

analysis, considering the vertical distribution of lateral forces on the structure. 

In any story in which Yi is less than or equal to 0.1, the structure need not be investigated 

further for stability concerns. When the quantity Yi in a story exceeds 0.1, the analysis of the 

structure should explicitly consider the geometric nonlinearity introduced by P-D effects. Most 

linear dynamic analysis software packages have the ability to consider P-D effects automatically. 

For nonlinear analysis procedures, second-order effects should be considered directly in the 

analysis; the geometric stiffness of all elements and components subjected to axial forces should 

be included in the mathematical model. When Yi in a story exceeds 0.3, the structure shall be 

considered unstable, unless a detailed global stability capacity evaluation for the structure, 

considering P-D effects, is conducted in accordance with the guidelines of Appendix A. 

Commentary: P-D effects can have very significant impact on the ability of 

structures to resist collapse when subjected to strong ground shaking. When the 

non-dimensional quantity, Y, calculated in accordance with Equation 2-3 

significantly exceeds a value of about 0.1, the instantaneous stiffness of the 

structure can be significantly decreased, and can effectively become negative. If 

earthquake induced displacements are sufficiently large to create negative 

instantaneous stiffness, collapse is likely to occur. 

Analyses reported in FEMA-355F, State of the Art Report on Performance 

Prediction and Evaluation, included direct consideration of P-D effects in 

determining the ability of regular, well configured frames designed to modern 

code provisions to resist P-D-induced instability and P-D-induced collapse. For 

regular, well configured structures, it is believed that if the value of Y is 

maintained within the limits indicated in this section, P-D-induced instability is 

unlikely to occur. Values of Y greater than this limit suggest that instability due 

to P-D effects is possible. In such cases, the frame should be reconfigured to 

provide greater resistance to P-D-induced instability unless explicit evaluation of 

these effects using the detailed Performance Evaluation methods outlined in 

Appendix A are performed. 

The evaluation approach for P-D effects presented in this section appears 

similar to but differs substantially from that contained in FEMA-302, and in use 

in the building codes for many years. The approach contained in FEMA-302 and 

the building codes was an interim formulation. The research conducted in support 

of these Recommended Criteria indicates that this interim approach was not 

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meaningful. Some of the research performed in support of these Recommended 

Criteria included the explicit evaluation of P-D effects for buildings of varying 

heights, subjected to many different types of ground motion, and designed using 

different building code provisions . Using these and other parameters, several 

tens of thousands of nonlinear analyses were run to investigate P-D effects. A 

complete discussion of the analyses supporting these recommendations may be 

found in FEMA-355F. Extensive additional discussion on the issue of P-D effects 

and their importance in the response of structures at large interstory drifts is 

contained in FEMA-355C, State of the Art Report on Systems Performance. 

2.8.7 Multidirectional Excitation Effects 

Buildings should be designed for seismic forces incident from any horizontal direction. For 

regular buildings, seismic displacements and forces may be assumed to act nonconcurrently in 

the direction of each principal axis of the building. For buildings with plan irregularity and 

buildings in which one or more components form part of two or more intersecting frames, 

multidirectional excitation effects should be considered. Multidirectional effects on components 

should include both torsional and translational effects. 

The requirement that multidirectional (orthogonal) excitation effects be considered may be 

satisfied by designing frames for the forces and deformations associated with 100% of the 

seismic displacements in one horizontal direction plus the forces associated with 30% of the 

seismic displacements in the perpendicular horizontal direction. Alternatively, it is acceptable to 

use the square root of the sum of the squares (SRSS) to combine multidirectional effects where 

appropriate. 

2.8.8 Vertical Excitation 

The effects of vertical excitation on horizontal cantilevers and prestressed elements should be 

considered by static or dynamic response methods. Vertical earthquake shaking may be 

characterized by a spectrum with ordinates equal to 67% of those of the horizontal spectrum 

unless alternative vertical response spectra are developed using site-specific analysis. Vertical 

earthquake effects on other beams and column elements should be evaluated for adequacy to 

resist vertical earthquake forces, as specified in FEMA-302. 

Commentary: There is no evidence that response to vertical components of 

ground shaking has had any significant effect on the performance of steel 

moment-frame structures. Consequently, the effect of this response is not 

recommended for consideration in the performance evaluation of these buildings, 

except as required by the building code. 

Traditionally, vertical response spectra, when considered, have been taken as 

2/3 of the horizontal spectra developed for the site. While this is a reasonable 

approximation for most sites, vertical response spectra at near-field sites, located 

within a few kilometers of the zone of fault rupture can have substantially 

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stronger vertical response spectra than indicated by this rule. Development of 

site-specific response spectra is recommended when vertical response must be 

considered for buildings on such sites. 

2.9 Frame Design 

The following provisions supplement the parallel provisions contained in the 1997 AISC 

Seismic Provisions. 

2.9.1 Strength of Beams and Columns 

Multi-story frames should be designed with a strong-column-weak-beam configuration, to 

avoid the formation of single-story mechanisms. As a minimum, Equation 9-3 of the 1997 AISC 

Seismic Provisions should be satisfied. In the application of Equation 9-3, the quantity Mc as 

defined in Section 3.2.6 of these Recommended Criteria should be substituted for the quantity 

M*

pb. 

Commentary: When subjected to strong ground shaking, multi-story structures 

with columns that are weaker in flexure than the attached beams can form singlestory mechanisms, in which plastic hinges form at the base and top of all columns 

in a story. Once such a mechanism forms in a structure, nearly all of the 

earthquake-induced lateral displacement will occur within the yielded story, 

which can lead to very large local drifts and the onset of P-D instability and 

collapse. 

Although weak-column-strong-beam designs are not desirable, the 1997 AISC 

Seismic Provisions does permit their use under certain conditions, even for 

Special Moment Frames. Before utilizing weak-column-strong-beam 

configurations, designers should be aware that the prequalified connections for 

Special Moment Frames contained in these Recommended Criteria are based on 

tests using strong columns. When considering moment frames which include 

columns deployed in the weak direction, designers should be aware that only one 

connection type (RBS, Section 3.5.5) has been tested for use with weak-direction 

columns for application in Special Moment Frames and, although those tests were 

successful, insufficient data exists to prequalify such connections. 

Nonlinear analyses of representative frames have clearly shown that the use 

of the provisions described above will not completely prevent plastic hinging of 

columns. This is because the point of inflection in the column may move away 

from the assumed location at the column mid-height once inelastic beam hinging 

occurs, and because of global bending induced by the deflected shape of the 

building, of which the column is a part. 

Except for the case when a column hinge mechanism forms, column hinging is 

not a significant problem, provided that the columns are designed as compact 

sections, are properly braced and axial loads are not high. It is well understood 

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that a column hinge will form at the base of columns which are continuous into a 

basement, or which are rigidly attached to a stiff and strong foundation. 

2.9.2 Lateral Bracing of Column Flanges 

Lateral bracing of column flanges, at beam-column connections should be provided whenever 

the following equation is not satisfied: 

*

� M 

pc 

� Mc 

‡ 2.0 (2-4) 

where: 

* M pc is the quantity defined in Section 9.6 of the 1997 AISC Seismic Provisions 

Mc is calculated as indicated in Section 3.2.6 of these Recommended Criteria. 

Commentary: The relationship indicated in Equation 2-4 has been included in 

proposals for the 2000 NEHRP Recommended Provisions for New Buildings (now 

under consideration by the Building Seismic Safety Council) as a trigger for 

requirements for lateral bracing of column flanges. Large axial loads reduce the 

ductility of column hinges. Consideration should be given to applying larger 

factors for columns with axial loads exceeding 50% of the critical column load. 

Bracing of the column at the location of the beam top flanges is normally 

supplied by the interconnection of the concrete slab, where such a slab occurs. At 

the location of the beam bottom flanges, sufficient lateral bracing can sometimes 

be shown to be provided by perpendicular beams and connected stiffeners for 

shallow column sections with wide flanges. Deeper beam-type sections, when 

used as columns, are typically less stable and normally will require direct lateral 

bracing of the flanges. See Section 2.9.6 for further guidelines on use of deep 

sections as columns. 

2.9.3 Panel Zone Strength 

Panel zones should conform to the strength requirements of Section 3.3.3.2 of these 

Recommended Criteria and the requirements of the individual prequalified connection design 

procedures. 

Commentary: Connection performance can be affected either positively or 

negatively by panel zone strength. Some shear yielding of the panel zone can 

relieve the amount of plastic deformation that must be accommodated in other 

regions of the frame and many connections have been found to provide the largest 

inelastic deformation capacity when yielding is balanced between the panel zone 

and other connection elements. However, excessive panel zone deformation can 

induce large secondary stresses into the connection that can degrade connection 

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performance and increase fracture toughness demand on welded joints, and can 

also cause deformations which are undesirable for column performance. For this 

reason, the individual connection prequalifications include limitations on panel 

zone strength relative to beam strength. 

2.9.4 Section Compactness Requirements 

Beams should conform to the section compactness requirements of AISC Seismic Provisions. 

Columns should also be compact, unless it can be shown by nonlinear analysis that the columns 

will not yield in response to the design earthquake. 

Commentary: The 1997 AISC Seismic Provisions provide section compactness 

requirements for beams used in moment frames, and for columns which may be 

subjected to hinging. The effect of beam flange b/t as it relates to connection 

performance is discussed in Section 3.3.1.1 Beam Flange Stability. The effect of 

beam section compactness on overall frame performance is directly related to 

how local buckling affects strength degradation of individual beams and columns 

in the frame. Flange local buckling and lateral torsional buckling are sources of 

strength degradation. 

It should be noted that for Reduced Beam Section (RBS) connections, the b/t in 

the area of beam hinging is reduced by the flange reduction, thereby reducing the 

propensity for flange local buckling. This may justify use of sections which are 

otherwise non-compact in frames employing these connections. See Section 

3.3.1.1 for recommendations. 

2.9.5 Beam Lateral Bracing 

The 1997 AISC Seismic Provisions require bracing of flanges of beams for Special Moment 

Frame systems. The unbraced length between supports is not permitted to exceed the quantity 

2500 ry /Fy. In addition, lateral supports are required where analysis indicates that a plastic hinge 

will form during inelastic deformations of the Special Moment Frame. General bracing of 

Special Moment Frame beams should conform to the AISC requirements. For bracing of beams 

at plastic hinges, refer to Section 3.3.1.5. 

2.9.6 Deep Columns 

The prequalified connections included in Chapter 3 of these Recommended Criteria are not 

prequalified for use with deep (beam-type) sections used as columns. The prequalified 

connections should only be used with W12 and W14 column sections. 

Commentary: Nearly all of the beam-column connection assemblies tested as 

part of this project, as well as by other researchers, utilized W14 column sections. 

In recognition of the fact that some designers prefer to use W24 or other deep 

section columns in order to increase frame stiffness economically, two tests of 

reduced beam section assemblies with W24 columns were conducted. These 

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assemblies performed poorly and one column failed through development of a 

fracture between the column web and flange. This fracture resulted from the 

combined effects of local torsional instability of the column and the presence of 

low-toughness material at the flange-to-web region, sometimes referred to as the 

k-area. The problem of low toughness material at the k-area of rolled structural 

shapes is a well documented phenomena related to the straightening practice used 

by some mills for certain ranges of shape. Additional information on this 

phenomena may be found in FEMA-355A, State of the Art Report on Base 

Materials and Fracture. However, there is relatively little data available on the 

instability of deep section columns in moment-resisting connections and this 

project was not able to develop adequate data on this effect to allow 

prequalification of connections with deep columns. 

2.9.7 Built-Up Sections 

The prequalified connections included in Chapter 3 of these Recommended Criteria have not 

been tested with built-up sections used as beams or columns. The prequalified connections 

should only be used with such sections when it can be shown that the built-up section conforms 

with all of the requirements for rolled sections as specified for the connection to be used. Of 

particular concern should be the strength of the web-to flange connection of the built-up section. 

2.10 Connection Design 

Chapter 3 of these Recommended Criteria provides criteria for the design and detailing of 

several types of prequalified Fully Restrained (FR) and Partially Restrained (PR) connections. 

These prequalified connections are recommended as acceptable for use in steel moment-frame 

systems, within the limitations expressed in Chapter 3, without further qualification analyses or 

tests. Table 2-2 lists the prequalified connection details, and the systems for which they are 

prequalified. All of these prequalifications apply only to frames composed of wide flange beams 

connected to the major axis of wide flange columns. 

In addition to the connections indicated in Table 2-2, Chapter 3 also provides information on 

several types of proprietary connections. Proprietary connections have not been prequalified by 

this project for service in specific systems. Engineers interested in the applicability of 

proprietary connections should obtain qualification information from the licensor. 

For each connection in Table 2-2, a complete set of design criteria is presented in Chapter 3. 

Depending on the selected system type, the designer should select a suitable connection, then 

follow the design criteria to complete the design. Connections contained in Chapter 3 may be 

used in applications outside the indicated range of prequalification provided that a projectspecific qualification program is followed, as indicated in Section 3.9. 

Connection types not prequalified under the guidelines of Chapter 3 may also be used, 

subject to the project-specific qualification procedures. 

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Table 2-2 Prequalified Connection Details 

Category Connection Description Acronym Permissible Systems 

Welded, fully 

restrained 

Welded Unreinforced Flanges, Bolted Web WUF-B OMF 

Welded Unreinforced Flanges, Welded Web WUF-W OMF, SMF 

Free Flange FF OMF, SMF 

Welded Flange Plate WFP OMF, SMF 

Reduced Beam Section RBS OMF, SMF 

Bolted, fully 

restrained 

Bolted, Unstiffened End Plate BUEP OMF, SMF 

Bolted, Stiffened End Plate BSEP OMF, SMF 

Bolted Flange Plates BFP OMF, SMF 

Bolted, partially 

restrained 

Double Split Tee DST OMF, SMF 

Commentary: For each of the prequalified connection types indicated in Table 2-

2, sufficient laboratory testing, together with related analytical work, has been 

performed to provide an ability to predict with confidence the limiting modes of 

behavior for the connection when properly constructed and the probability that 

the connection will be able to sustain certain levels of inelastic deformation. This 

confidence only applies to application within certain limits, including material 

specifications, and member sizes. If a design falls outside the range of 

prequalification for a connection detail, it is necessary to extend the existing 

qualification for use in the specific application, by performing additional 

laboratory prototype testing. Chapter 3 indicates the extent of the additional 

testing recommended to extend connection qualification, on a project-specific 

basis, as well as more general recommendations for prequalifying connection 

details for broader application. 

2.11 Specifications 

FEMA-353 – Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications presents supplemental recommendations 

for fabrication and erection of steel moment-frame structures. These supplemental 

recommendations address welding and base materials, methods of fabrication and quality 

assurance. It is recommended that project specifications include the specific paragraphs of 

FEMA-353 that are applicable to the design being used. 

Commentary: FEMA-353 is written in the form of supplemental provisions to the 

existing provisions of the building codes, FEMA-302, and standard AISC, AWS 

and ASTM specifications. It is expected that eventually, these standard 

specifications and provisions will be amended to adopt the supplemental 

provisions recommended by FEMA-353. In the interim, the applicable sections 

and paragraphs can be reproduced in individual project specifications. When this 

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is done, it is recommended that the specific language taken from the reference be 

used without modification and attributed to the source, so that fabricators and 

erectors can readily recognize and become accustomed to the use of the FEMA353 requirements. 

2.12 Quality Control and Quality Assurance 

FEMA-353 – Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications provides complete guidelines and 

commentary for Quality Control and Quality Assurance. The designer should utilize those 

guidelines to ensure the proper selection and handling of materials and shop and field fabrication 

of moment-frame connections. 

Commentary: FEMA-353 has a complete discussion of quality control 

recommendations and the reasons for them. Quality control and quality 

assurance are important for the achievement of the intended performance. 

2.13 Other Structural Connections 

2.13.1 Column Splices 

Column splices in moment frames should be designed to develop the full bending and shear 

strength of the column, unless an inelastic analysis is performed to determine the largest axial 

loads, moments and shears likely to occur at the location of the splice and the splice detail can be 

shown to be adequate to resist these axial loads, moments and shears, considering stress 

concentrations inherent in the types of joints being used. 

Welded flange splices may be made either with full penetration groove welds, or with splice 

plates fillet welded to the column flanges. Weld metal with a minimum rated toughness as 

described in Section 3.3.2.4 should be used and weld tabs should be removed. Bolted column 

flange splices should be designed to preclude net section fracture, block shear failure, and bolt 

pull-through failure of the column flange or of the splice plates. 

Column web splices may be either bolted or welded, or welded to one column piece and 

bolted to the other. Bolted splices using plates or channels on both sides of the column web are 

preferred because of the inherent extra safety afforded by “capturing” the web. Partial Joint 

penetration welded web splices are not recommended. Column web splices should be designed 

to resist the maximum shear force that the column is capable of producing. 

Splices of columns that are not a part of the seismic-force-resisting system should be made in 

the center one-third of the column height, and should have sufficient shear capacity in both 

orthogonal directions to maintain the alignment of the column at the maximum shear force that 

the column is capable of producing. 

Commentary: Section 8.3 of the 1997 AISC Seismic Provisions specifies 

requirements for design of column splices for columns that are part of the 

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seismic-force-resisting system. The requirements prohibit splices made with fillet 

welds or partial penetration groove welds located within four feet or within onehalf the column clear height of the beam-to-column connections. This prohibition 

is because fillet welds in tensile applications and partial penetration butt welds 

are both details with relatively low tensile capacity and poor inelastic capability. 

For typical cases, the prohibition against such splices within four feet of a beamcolumn joint will control. The one-half column height requirement is intended to 

apply to those rare cases when the clear column height is less than eight feet. The 

1997 AISC Seismic Provisions permit such splices in the mid-height zone of 

columns based on the belief that large flexural demands, and in particular 

inelastic demands are unlikely to occur in this region. Inelastic analyses of 

frames, however, clearly demonstrate that this presumption is incorrect for 

frames subjected to seismic loadings that exceed their elastic capacity. For this 

reason, as well as the severe potential consequences of column splice failure, the 

1997 AISC Seismic Provisions are not considered to be sufficiently conservative 

in this area. 

Because bending and axial stresses at column splice welds may be high, it is 

recommended that weld filler metals with rated notch toughness be used for these 

splices and that runoff tabs be removed. Where CJP welds are used, removal of 

backing is not judged to be necessary because the configuration of backing for 

column-to-column flange welds is not conducive to crack formation, as it is for 

the right-angle condition of beam-to-column flange joints. Properly designed 

bolted flange splices may be shown to be adequate for some column splice 

applications. 

Bolted web connections are preferred by many engineers and contractors 

because they have advantages for erection, and, when plates are placed on both 

sides of the web, they are expected to maintain alignment of the column in the 

event of a flange splice fracture. Partial joint penetration welded webs are not 

recommended, because fracture of a flange splice would likely lead to fracture of 

the web splice, considering the stress concentrations inherent in such welded 

joints. 

Inelastic analyses have shown the importance of the columns that are not part 

of the seismic-force-resisting system in helping to distribute the seismic shears 

between the floors. Even columns that have beam connections that act as pinned 

connections may develop large bending moments and shears due to non-uniform 

drifts of adjacent levels. For this reason, it is considered to be important that 

splices of such columns be adequate to develop the shear forces corresponding to 

development of plastic hinges at the ends of the columns in both orthogonal 

directions. 

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2.13.2 Column Bases 

Column bases can be of several different types, as follows: 

1. The column may continue into a basement, crawl space, or grade beam, in such a way that the 

column’s fixity is assured without the need for a rigid base plate connection. 

2. Large columns may be provided at the bottom level to limit the drift, and a “pinned base” 

may be utilized. 

3. A connection which provides partial fixity may be provided, so that the column base is fixed 

up to some column moment, but the base itself yields before the column hinges. 

4. A heavy base plate assembly may be provided which is strong enough to force yielding in the 

column. 

In all of these cases, the designer should consider the base connection as similar to a beam-tocolumn connection and apply similar principles of design and detailing. 

Notes: 

1. For the first case above, the designer should recognize that hinging will occur in the column, 

just above the first floor. The horizontal shear to be resisted at the ends of the column in the 

basement level should be calculated considering the probable overstrength of the framing. 

2. For the “pinned base”, the designer should ensure that the required shear capacity of the base 

can be maintained up to the maximum rotation that may occur. 

3. In designing a base with partial fixity , the designer should consider the principles used in the 

design of partially-restrained connections. This type of base may rely on bending of the base 

plate (similar to an end plate connection), bending of angles or tees, or yielding of anchor 

bolts. In the latter case, it is necessary to provide bolts or rods with adequate elongation 

capacity to permit the required rotation and sufficient unrestrained length for the yielding to 

occur. Shear capacity of the base plate to foundation connection must be assured at the 

maximum rotation. 

4. For the fully fixed base, the designer should employ the same guidelines as given for the rigid 

fully-restrained connections. Such connections may employ thick base plates, haunches, 

cover plates, or other strengthening as required to develop the column hinge. Where 

haunched type connections are used, it must be recognized that the hinging will occur above 

the haunch, and appropriate consideration should be given to the stability of the column 

section at the hinge. 

Commentary: It is well recognized that achievement of a mechanism in a 

moment frame requires a hinge at, or near to, the base of the column. The column 

base detail must accommodate the required hinging rotations while maintaining 

the strength required to provide the mechanism envisioned by the designer. These 

conditions are similar to the requirements for beam-to-column connections, as 

described. 

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2.13.3 Welded Collectors and Chords 

Connections of highly loaded collectors and chords are often made with welded or bolted 

flange details comparable to those employed in moment frames. Design of such connections 

should incorporate the principles applied to moment-frame connections, unless it can be shown 

that the connection will remain elastic under the combination of the axial load, calculated at the 

limit strength of the system, and the corresponding rotation due to building drift. 

Commentary: The rotational demand on rigid connections made for other 

purposes are often comparable to those of moment-frame beams. When coupled 

with high axial loads, demands on welded or bolted joints can be high. The 

principles of design for moment-frame beam-to-column connections are 

applicable to such conditions. 

2.13.4 Simple Beam-to-Column Gravity Connections 

Simple welded shear tab connections of beams to columns in buildings employing moment 

frames and other relatively flexible lateral-force-resisting systems should utilize details that have 

been demonstrated to have sufficient rotational capacity to accommodate the rotations that occur 

at the anticipated drifts, while maintaining capacity for the required gravity forces. In the absence 

of a more detailed analysis, adequate rotation capacity can be considered to be that associated 

with the design story drift calculated using the methods of FEMA-302 multiplied by 1.5. As 

described in the commentary below, calculations to justify the adequacy of this condition should 

not be necessary under normal conditions. 

When deep beams with deep bolt groups are connected to small columns, the columns should 

be compact, or sufficient rotational capacity should be provided in the connections to preclude 

hinging of the column when subjected to the drift calculated as described above. 

Commentary: Research conducted under this project has shown that the plastic 

rotational capacity of simple bolted shear tab type connections, designed using 

the methods of the AISC LRFD Specification, and with adequate clearance of 

beam flanges from the column flanges to prevent bearing, is dependent on the 

depth of the bolt group, dbg, and can reasonably be calculated as: 

q p = 0.15 - 0.0036dbg (2-5) 

where dbg is the vertical dimension of the bolt group in inches. The additional 

elastic rotational capacity of these connections is estimated as about 0.02 

radians. This gives a total estimated drift capacity for such connections of: 

q p = 0.17 - 0.0036dbg (2-6) 

The use of Equation 2-6 above will result in a calculated rotational capacity 

of more than 0.09 radian for an 8-bolt group with bolts spaced at 3”, which will 

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Chapter 2: General Requirements Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

be more than adequate for most conditions. Where the calculated rotational 

angle is not sufficient, slotted holes in the shear tab, or other means of 

accommodating larger rotations should be used. It should be noted that rotation 

capacities for connections made with clip angles bolted to the beam have not been 

found to be significantly higher than those for welded shear tabs. Refer to FEMA355D for additional information. 

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Moment-Frame Buildings Chapter 3: Connection Qualification

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3. CONNECTION QUALIFICATION 

3.1 Scope 

This chapter provides design procedures and qualification data for various types of 

connections for new steel moment-frame buildings. Included herein are criteria for design of 

connections and conditions that are generic to most connection upgrade types, and criteria for 

specific details of connections intended to be prequalified for use in seismic applications. Each 

of the connection prequalifications is limited to specific conditions for which they are applicable, 

including member size ranges, grades of material and other details of the connection. Also 

included in this chapter are recommended criteria for qualification of connections that have not 

been prequalified or are proposed for use outside the limits of their prequalification, as set forth 

herein, and information on several types of proprietary connections. 

Commentary: The 1988 Uniform Building Code (ICBO, 1988) introduced a single 

prequalified (“prescriptive”) moment-connection design for seismic applications, 

representative of prevailing west coast practice at the time. The “qualification” 

of this connection was based primarily on the research of Popov and Stephen in 

the early 1970s. The UBC prequalified connection was subsequently adopted into 

the 1992 AISC Seismic Provisions and then into other model building codes. 

The 1994 Northridge earthquake demonstrated that this prescriptive 

connection, as it was being used in contemporary practice, was inadequate for the 

anticipated seismic demands. Following this discovery, enforcement agencies 

adopted emergency changes to the building codes, deleting the prescriptive 

connection and requiring that all connection details used in moment resisting 

frames for seismic application be qualified for adequacy through a program of 

prototype testing. The Interim Guidelines for Inspection, Evaluation, Repair, 

Modification and Design of Welded Moment-Resisting Steel Frames (FEMA-267) 

and, the companion Interim Guidelines Advisories (FEMA-267A and FEMA267B), continued and reinforced the recommendation for permitting the use of 

only those connection details demonstrated as adequate by a program of 

prototype testing, while providing extensive guidance on how and under what 

conditions such testing should be required and how test results might be 

interpolated or extrapolated. These recommendations were adopted with some 

modification, by FEMA-302, the 1997 AISC Seismic Provisions, and the Uniform 

Building Code (ICBO, 1997), which require that connections for all types of 

moment frames be qualified by test. Connections for Ordinary Moment Frames 

(OMFs) were permitted to be designed based on calculations alone, if certain 

strength and detailing conditions were met. 

In the time since the publication of those documents more than 150 connection 

assemblies have been tested, allowing new prequalifications for connection 

details believed to be capable of providing reliable service to be developed. 

Those prequalifications applicable to the design of new structures appear in these 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

Recommended Criteria. It is the intent of these criteria to return the design of 

steel moment-frame structures to the straightforward select-design-detail task, 

while providing the reliability that was previously incorrectly assumed to exist. 

For the majority of structures and conditions of use, it is intended that the 

designer will be able to select, design, and detail prequalified moment-frame 

connections appropriate for the intended structure by using the criteria of this 

chapter, without the need to perform project-specific prototype qualification 

testing. For connection details other than those included herein, prototype 

qualification testing must still be performed, and recommended criteria are 

provided for performance and acceptance of such testing. 

The research supporting the connection prequalifications contained in this 

chapter is summarized in FEMA-355D, State of the Art Report on Connection 

Performance. The interested reader is referred to that report for more 

background on these recommendations, including complete references to specific 

research reports where more extensive descriptions of individual research 

methods and results can be found. 

3.2 Basic Design Approach 

This section provides recommended criteria on basic principles of connection design, 

including selection of an appropriate connection type, estimation of locations of inelastic 

behavior (formation of plastic hinges), determination of probable plastic moment at the plastic 

hinges, determination of shear at the plastic hinge, and determination of design strength demands 

at critical sections of the assembly. These basic principles apply to the recommended calculation 

procedures for all prequalified connection types. 

3.2.1 Frame Configuration 

Frames should be proportioned and detailed so that the required interstory drift angle for the 

frame can be accommodated through a combination of elastic deformation and the development 

of plastic hinges at pre-determined locations within the frame. Figure 3-1 indicates a frame in 

which inelastic drift is accommodated through the development of plastic flexural deformation 

(plastic hinges) within the beam span, remote from the face of the column. Such behavior may 

be obtained by locally stiffening and strengthening fully restrained connections by using cover 

plates, haunches and similar detailing, such that the ratio of flexural demand to plastic section 

capacity is maximum at these interior span locations. This condition can also be obtained by 

locally reducing the section of the beam at desired locations for plastic hinging to obtain a 

condition of maximum flexural demand to plastic section capacity at these sections. Other 

locations where plastic deformation may take place in frames, depending on the configuration, 

detailing and relative strength of the beams, columns and connections include: within the 

connection assembly itself, as is common for partially restrained connections; within the column 

panel zone; or within the column. The total interstory drift angle, as used in these criteria is equal 

to the sum of the plastic drift, as described here, and that portion of the elastic interstory drift 

resulting from flexural deformation of the individual members. Interstory drift resulting from 

axial deformations of columns is not included. 

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Moment-Frame Buildings Chapter 3: Connection Qualification

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Plastic 

hinges 

Deformed 

frame shape 

Undeformed 

frame 

L’ 

L 

h 

Drift angle − q 

Figure 3-1 Inelastic Behavior of Frames with Hinges in Beam Span 

Commentary: Nonlinear deformation of frame structures is accommodated 

through the development of inelastic flexural or shear strains within discrete 

regions of the structure. At large inelastic strains these regions can develop into 

plastic hinges that can accommodate significant concentrated rotations at 

constant (or nearly constant) load through yielding at tensile fibers and yielding 

and buckling at compressive fibers. If a sufficient number of plastic hinges 

develop in a frame, a mechanism is formed and the frame can deform laterally in 

a plastic manner. This behavior is accompanied by significant energy dissipation 

and potentially substantial damage to the highly strained elements. The 

formation of hinges in columns, as opposed to beams, is undesirable, as this may 

result in the formation of mechanisms with relatively few elements participating, 

so called “story mechanisms,” and consequently little energy dissipation 

throughout the structure. 

The prequalified connection contained in the building codes prior to the 1994 

Northridge earthquake was presumed to result in a plastic behavior that consisted 

of development of plastic hinges within the beams at the face of the column, or 

within the column panel zone, or as a combination of the two. If the plastic hinge 

develops primarily in the column panel zone, the resulting column deformation 

may result in very large secondary stresses on the beam flange to column flange 

joint, a condition that, for certain types of connections, can contribute to brittle 

failure. If the plastic hinge forms in the beam at the face of the column, this can 

result in large inelastic strain demands on the weld metal and surrounding heataffected zones. These conditions can lead to brittle failure. 

Special Moment Frame (SMF) structures are expected to be capable of 

extensive amounts of energy dissipation through the development of plastic 

hinges. In order to achieve reliable performance of these structures, frame 

configurations should incorporate a strong-column-weak-beam design that can 

lead to development of column hinging and story collapse mechanisms. Further, 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

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fully restrained beam-column connections should be configured either to force the 

inelastic action (plastic hinge) away from the column face, where performance is 

less dependent on the material and workmanship of the welded joint, or must 

employ optimum welded joint design and quality assurance measures. Shifting 

the hinge away from the column face can be done either by local reinforcement of 

the connection, or by locally reducing the cross section of the beam at a distance 

away from the connection. Plastic hinges in steel beams have finite length, 

typically on the order of half the beam depth. Therefore, for this approach, the 

location for the plastic hinge should be shifted at least that distance away from 

the face of the column. For situations where unreinforced connections employing 

optimum joint design, fabrication, and quality assurance are used, the plastic 

hinges will occur about one-quarter of the beam depth from the column face and 

will extend to the face of the column. When the plastic hinge location is shifted 

away from the face of the column, the flexural demands on the columns, for a 

given beam size, are increased. Care must be taken to ensure that weak column 

conditions are not inadvertently created by local strengthening of the connections. 

Connection configurations of the type described above, while believed to be 

effective in preventing brittle connection fractures, will not prevent structural 

damage from occurring. Brittle connection fractures are undesirable for several 

reasons. First, severe connection degradation can result in loss of gravity load 

carrying capacity of the framing at the connection and the potential development 

of local collapse. From a global perspective, the occurrence of many connection 

fractures results in a substantial reduction in the lateral-force-resisting strength 

and stiffness of the structure which, in extreme cases, can result in instability and 

collapse. Connections configured as described in these Recommended Criteria 

should experience fewer such brittle fractures than unmodified connections. 

However, the formation of a plastic hinge within the beam is not a completely 

benign event. Beams that have experienced significant plastic rotation at such 

hinges may exhibit large buckling and yielding deformation, as well as localized 

damage to floor slabs and other supported elements. In severe cases, this damage 

must be repaired. The cost and difficulty of such repairs could be comparable to 

the costs incurred in repairing fracture damage of the type experienced in the 

Northridge earthquake. The primary difference is that life safety protection will 

be significantly enhanced and most structures that have experienced such plastic 

deformation damage should continue to be safe for occupancy, while repairs are 

made. 

If the types of damage described above are unacceptable for a given building, 

then alternative structural systems, which will reduce the plastic deformation 

demands on the structure during a strong earthquake, should be considered. 

Appropriate methods of achieving such goals include the installation of 

supplemental braced frames, energy dissipation systems, base isolation systems 

and similar structural systems. Framing systems incorporating partially 

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Moment-Frame Buildings Chapter 3: Connection Qualification

Criteria for New Steel FEMA-350 

restrained connections may also be effective in resisting large earthquake induced 

deformation with limited damage. 

Ordinary Moment Frame structures are designed so that they will experience 

less inelastic deformation than Special Moment Frame structures for a given 

ground motion. Therefore, for Ordinary Moment Frame systems, fully restrained 

connections that permit development of plastic hinges at locations other than 

within the beam span, e.g. in the panel zone or in the column, are permitted. 

Partially restrained connections are configured to form plastic hinges through 

yielding of the connection elements themselves. The plastic moment capacity of 

these connections is typically a fraction of that of the connected framing elements, 

encouraging the inelastic behavior to occur within the connection at relatively 

low force levels. These connections must be configured to ensure that inelastic 

behavior occurs through ductile yielding of elements, rather than brittle failure, 

such as shearing or elongation of bolts, or tensile fractures through weak netsections of connection elements. Frames employing properly designed partially 

restrained connections can be capable of extensive inelastic response, with plastic 

hinges forming within the connection, adjacent to the face of the column. Because 

such connections are weaker and less stiff, systems using partially restrained 

connections typically incorporate more of the framing members into the momentframe system than do frames using fully restrained connections. 

3.2.2 Connection Configuration 

A connection configuration should be selected that is compatible with the selected structural 

system and the sizes of the framing elements. Sections 3.5 and 3.6 present data on a series of 

prequalified connections, from which an appropriate connection type may be selected. 

Alternatively, if project-specific connection qualification in accordance with Section 3.9 is to be 

performed, a connection of any configuration that provides the appropriate interstory drift 

capacity, in accordance with Section 3.9.2 and meets the strength and stiffness demands for the 

structure, may be selected. 

3.2.3 Determine Plastic Hinge Locations 

Based on the data presented in Tables 3-2 through 3-6 and 3-8 through 3-12 for prequalified 

connections, or data obtained from a qualification testing program for configurations that are 

qualified on a project-specific basis, the location of expected plastic hinge formation sh as 

indicated in Figure 3-2 should be identified. The plastic hinge locations presented for 

prequalified connections are valid for beams with gravity loads representing a small portion of 

the total flexural demand. For frames in which gravity loading produces significant flexural 

stresses in the members, locations of plastic hinge formation should be determined based on 

methods of plastic analysis. 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

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ıShˆ 

Beam depth − d 

L’ 

Plastic 

hinge 

Reduced beam 

section 

(if applicable) 

ıSh ˆ 

Connection 

reinforcement 

(if applicable) 

L 

Figure 3-2 Location of Plastic Hinge Formation 

Commentary: The suggested location for the plastic hinge, as indicated by the 

parameter sh in the prequalification data, is valid only for frames with limited 

gravity loading present on the frame beams. If significant gravity load is present, 

this can shift the locations of the plastic hinges, and in the extreme case, even 

change the form of the collapse mechanism. If flexural demand on the girder due 

to gravity load is less than about 30% of the girder plastic capacity, this effect 

can safely be neglected, and the plastic hinge locations taken as indicated. If 

gravity demands significantly exceed this level then plastic analysis of the frame 

should be performed to determine the appropriate hinge locations. 

3.2.4 Determine Probable Plastic Moment at Hinges 

For fully restrained connections designed to develop plastic hinging in the beam or girder, the 

probable plastic moment at the location of the plastic hinge should be determined as: 

M pr = CprRyZeFy (3-1) 

where: 

Mpr = probable peak plastic hinge moment, 

Cpr = a factor to account for the peak connection strength, including strain hardening, 

local restraint, additional reinforcement, and other connection conditions. For 

most connection types, Cpr is given by the formula: 

F +F 

Cpr = y u (3-2) 2 Fy 

A value of 1.2 may be used for all cases, except where otherwise noted in the 

individual connection design procedures included with the prequalifications in 

later sections of these Recommended Criteria. 

Ry = A coefficient, applicable to the beam or girder material, obtained from the 1997 

AISC Seismic Provisions. 

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Ze = The effective plastic modulus of the section (or connection) at the location of the 

plastic hinge. 

Fy = the specified minimum yield stress of the material of the yielding element. 

Fu = the specified minimum tensile stress of the material of the yielding element. 

For connections that do not develop plastic hinges in the beam, the hinge strength should be 

calculated, for the pertinent yield mechanism as confirmed by tests, considering the variation in 

material properties of the yielding elements. For prequalified connections, calculation methods 

to determine the yield strengths of the various active mechanisms are given in Sections 3.5, 3.6, 

and 3.7. 

Commentary: The 1997 AISC Seismic Provisions use the formulation 1.1RyMp for 

calculation of the expected plastic moment capacity of a beam As described in 

FEMA-355D, State of the Art Report on Connection Performance, research has 

shown that, for most connection types, the peak moment developed is somewhat 

higher than the 1.1 factor would indicate. Therefore, in these Recommended 

Criteria, the factor Cpr is used for individual connections, with a default value of 

1.2 applicable to most cases. 

3.2.5 Determine Shear at the Plastic Hinge 

The shear at the plastic hinge should be determined by methods of statics, considering gravity 

loads acting on the beam. A free body diagram of that portion of the beam between plastic 

hinges is a useful tool for obtaining the shear at each plastic hinge. Figure 3-3 provides an 

example of such a calculation. For the purposes of such calculations, gravity load should be 

based on the load combinations indicated in Section 3.4.1. 

3.2.6 Determine Strength Demands at Each Critical Section 

In order to complete the design of the connection, including, for example, sizing the various 

plates, bolts, and joining welds which make up the connection, it is necessary to determine the 

shear and flexural strength demands at each critical section. These demands may be calculated 

by taking a free body of that portion of the connection assembly located between the critical 

section and the plastic hinge. Figure 3-4 demonstrates this procedure for two critical sections, for 

the beam shown in Figure 3-3. 

Commentary: Each unique connection configuration may have different critical 

sections. The vertical plane that passes through the joint between the beam 

flanges and column (if such joining occurs) will typically define at least one such 

critical section, used for designing the joint of the beam flanges to the column. A 

second critical section occurs at the center line of the column. Moments 

calculated at this point are used to check strong-column-weak-beam and panel 

zone shear conditions. Other critical sections are described in the design 

procedures for each connection type. 

3-7
P

M pr M pr

Vp

V A

w

Vp

MfMpr

Critical Section at Column Face Critical Section at Column Centerline 

= M f pr Vx M c = pr + Vxd p ( 2)

Vp

Mpr

Vp

MfMpr

Critical Section at Column Face Critical Section at Column Centerline 

= M f pr Vx M c = pr + Vxd p ( 2)

Vp

Mpr

Vp

MfMpr

Vp

Mf Mpr

Critical Section at Column Face Critical Section at Column Centerline 

= M f pr Vx M c = pr + Vxd p ( 2)

Vp

Mpr

Vp

Mpr

Vp

M + p + M c M + p + M c M + p + M c

Recommended Seismic Design 

Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

L 

L sh ’ 

Plastic 

hinge P 

L/2 

P V A 

w 

M pr 

L’ 

ıAˆM pr 

Vp 

taking the sum of moments about ıAˆ = 0 

M + M + P L' 2 + W L'

2 

pr pr 2

Vp = L' 

Figure 3-3 Sample Calculation of Shear at Plastic Hinge 

M f 

Plastic 

hinge 

M pr 

x+dc/2 

dc

Mc

Plastic 

hinge 

M pr 

Vp Vp 

x 

Mf = Mpr +Vx p Mc = Mpr + Vxd p ( + c 2)

Critical Section at Column Face Critical Section at Column Centerline 

Figure 3-4 Calculation of Demands at Critical Sections 

3.2.7 Yield Moment 

The design procedures for some prequalified connections contained in these Recommended 

Criteria require that the moment at the face of the column at onset of plastic hinge formation, 

Myf, be determined. Myf may be determined from the following equation: 

M yf = CyM f (3-3) 

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Moment-Frame Buildings Chapter 3: Connection Qualification

Criteria for New Steel FEMA-350 

where: 

1 Cy = Z (3-4) 

C be 

pr Sb 

Cpr = the peak connection strength coefficient defined in Section 3.2.5 

Sb = the elastic section modulus of the beam at the zone of plastic hinging 

Zbe = the effective plastic section modulus of the beam at the zone of plastic hinging. 

3.3 General Requirements 

This section provides criteria for connection design conditions that are considered to be 

general, that is, those conditions which, when they occur in a connection, are considered to 

perform in a similar way, or at least to have the same requirements for successful performance, 

irrespective of the connection type being used. The designer should employ these criteria in the 

design of all connection types, except when specific testing has been performed that qualifies the 

connection for use with different conditions, or unless otherwise specifically indicated in these 

Recommended Criteria. 

3.3.1 Beams 

3.3.1.1 Beam Flange Stability 

Beam flange slenderness ratios bf/2tf (b/t) should be limited to a maximum value of 52/ �Fy, 

as required by the 1997 AISC Seismic Provisions. For moment frame beams with RBS 

connections, it is recommended that the bf /2tf be determined based on the flange width (bf) 

measured at the ends of the center 2/3 of the reduced section of beam unless gravity loads are 

large enough to shift the hinge point significantly from the center point of the reduced section. 

Commentary: The AISC Seismic Provisions require that beam flange slenderness 

ratios bf /2tf (b/t) be limited to a maximum of 52/ �Fy. This specific value is 

intended to allow some plastic rotation of the beam to occur before the onset of 

local buckling of the flanges, a highly undesirable phenomenon. Buckling of most 

of the beam flanges in a moment resisting frame results in development of frame 

strength degradation increasing both story drifts and the severity of P-D effects 

and therefore should be avoided. Local flange buckling results in large local 

straining of the flanges and the early on-set of low-cycle fatigue induced tearing 

of the beam flanges, which ultimately limits the ability of the assembly to 

withstand cyclic inelastic rotation demands. Further, severely buckled beam 

flanges can be even more difficult to repair than fractured beam connections. 

Notwithstanding the above, under large plastic rotation demands, buckling of 

beam flanges will inevitably occur. The value of the b/t of the beam involved in a 

specific connection can have a major effect on how the beam column assembly 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

performs. Beams and girders used in moment frames should comply with the 

limits specified by AISC, except as specifically modified by individual connection 

prequalifications or qualification tests. It should be noted that under this 

program, many assemblies with W30x99 beams conforming to ASTM A572 were 

tested. Although this section has bf /2tf equal to 54/�Fy they performed 

acceptably. 

3.3.1.2 Beam Web Stability 

Web height-to-thickness ratios, hc /tw for beams in moment resisting frames should not 

exceed 418/�Fy. 

Commentary: The 1997 AISC Seismic Provisions permit use of beams with web 

hc /tw ratios as high as 520/�Fy, for beams without axial load. Most of the testing 

conducted in support of the development of these Recommended Criteria utilized 

either W30x99 or W36x150 beam sections. Both of these structural shapes have 

hc /tw ratios that conform to the recommended 418/�Fy ratio, as do nearly all 

commonly rolled shapes. Since many of the specimens exhibited significant web 

buckling in the area of plastic hinges, it is not considered prudent to utilize beams 

with thinner webs in moment resisting frames. Although stiffening of the webs 

could be done to limit web buckling, it is possible that such stiffeners could be 

detrimental to connection performance. Since connections with web stiffeners 

were not tested, such connections have not been prequalified. Refer to FEMA355D, State of the Art Report on Connection Performance, for further discussion 

of web buckling of moment-frame beams. 

3.3.1.3 Beam Depth and Span Effects 

The prequalified connections contained in Sections 3.5, 3.6, and 3.7 of these Recommended 

Criteria are limited in application to specific beam depths and span-to-depth ratios. These 

limitations are noted in the tabulated data for each connection. For frames designed using 

project-specific connection qualifications, connection tests used in the connection qualification 

program should employ beams of similar or greater depth than those used in the frame and 

similar or smaller span-to-depth ratio. 

Commentary: Both beam depth and beam span-to-depth ratio are significant in 

the inelastic behavior of beam-column connections. At a given induced curvature, 

deep beams will undergo greater straining than shallower beams. Similarly, 

beams with shorter span-to-depth ratio will have a sharper moment gradient 

across the beam span, resulting in reduced length of the beam participating in 

plastic hinging and increased strains under inelastic rotation demands. Most of 

the beam-column assemblies tested under this project used configurations 

approximating beam spans of about 25 feet and beam depths varying from W30 to 

W36 so that beam span-to-depth ratios were typically in the range of 8 to 10. 

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Moment-Frame Buildings Chapter 3: Connection Qualification

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Additional information may be found in FEMA-355D, State of the Art Report on 

Connection Performance. 

3.3.1.4 Beam Flange Thickness Effects 

The prequalified connections contained in Sections 3.5, 3.6, and 3.7 of these Recommended 

Criteria are limited in application to specific beam flange thicknesses. These limitations are 

noted in the tabulated data for each connection. For frames designed using project-specific 

connection qualifications, connection tests used in the connection qualification program should 

employ beam flanges of similar or greater thickness than those used in the frame. 

Commentary: In addition to controlling the stability of the flange under 

compressive loading, as described in Section 3.3.1.1, beam flange thickness also 

affects the size of welds in welded connections. Although it is not a given that 

larger welds will be less reliable than smaller welds, greater control may be 

necessary to assure their performance, and quality control may be more difficult. 

Additionally, residual stresses are likely to be higher in thicker material with 

thicker welds. 

3.3.1.5 Lateral Bracing at Beam Flanges at Plastic Hinges 

Plastic hinge locations that are remote from the column face in beams that do not support a 

slab should be provided with supplemental bracing, as required by the 1997 AISC Seismic 

Provisions. Where the beam supports a slab and is in direct contact with the slab along its span 

length, supplemental bracing need not be provided. 

Commentary: The 1997 AISC Seismic Provisions require that beam flanges be 

braced at plastic hinge locations. Because plastic hinges have been moved away 

from the column face for some of the connection types in this section, a strict 

interpretation of the provisions would lead to a requirement that flanges at such 

hinges be laterally braced. Limited testing conducted as part of this project 

(FEMA-355D) suggests that, as long as the hinging beam is connected to a 

concrete slab, excessive strength deterioration due to lateral buckling will not 

occur within the ranges of drift angle normally considered important. Therefore, 

these Recommended Criteria do not require supplemental bracing of plastic hinge 

locations adjacent to column connections of beams supporting slabs. 

For those cases where supplemental bracing of beam flanges near plastic 

hinges is appropriate, care must be taken in detailing and installation of bracing 

to assure that detrimental attachments are not made directly within the area of 

anticipated plastic behavior. This is because of the inherent risk of reducing 

plastic deformation capacity for the beam by introducing stress concentrations or 

metallurgical notches into the region of the beam that must undergo plastic 

straining. See FEMA-355D, State of the Art Report on Connection Performance, 

for further discussion of flange bracing. 

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FEMA-350 Criteria for New Steel 

3.3.1.6 Welded Shear Studs 

Welded shear studs, or other attachments for composite action with slabs or for diaphragm 

shear transfer, should not be installed within the hinging area of moment frame beams. The 

hinging area is defined as the distance from the column flange face to one half the beam depth 

beyond the theoretical hinge point. Shot-in, or screwed attachments should not be permitted in 

this area either. 

Commentary: It has been shown in some tests that welded shear studs and the 

rapid increase of section caused by composite action can lead to beam flange 

fractures when they occur in the area of the beam flange that is undergoing large 

cyclic strains. It is not certain whether the welding of the studs, the composite 

action, or a combination of the two is the cause, but, based on the limited 

evidence, it is judged to be prudent to permit no studs in the hinging area. It is 

also prudent to permit no attachments, which involve penetration of the flanges in 

the hinging region. 

3.3.2 Welded Joints 

3.3.2.1 Through-Thickness Strength 

The through-thickness strength of column material conforming to FEMA-353, Recommended 

Specifications and Quality Assurance Guidelines for Steel Moment-Frame Construction for 

Seismic Applications, need not be explicitly checked in connection design, except where required 

by the design procedures for a specific prequalified connection. 

Commentary: Early investigations of connection fractures in the 1994 Northridge 

earthquake identified a number of fractures that appeared to be the result of 

inadequate through-thickness strength of the column flange material. As a result 

of this, in the period immediately following the Northridge earthquake, a number 

of recommendations were promulgated that suggested limiting the value of 

through-thickness stress demand on column flanges to a value of 40 ksi, applied 

to the projected area of the beam flange attachment. This value was selected to 

ensure that through-thickness yielding did not initiate in the column flanges of 

fully restrained connections and often controlled the overall design of a 

connection subassembly. 

It is important to prevent the inelastic behavior of connections from being 

controlled by through-thickness yielding of column flanges. This is because it 

would be necessary to develop very large local ductilities in the column flange 

material in order to accommodate even modest plastic rotation demands on the 

assembly. However, the actual cause for the fractures that were initially 

identified as through-thickness failures of the column flange are now believed to 

be unrelated to this material property. Rather, it appears that this damage 

occurred when fractures initiated in defects present in the CJP weld root, not in 

the flange material (FEMA-355E). These defects sometimes initiated a crack, that 

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under certain conditions, propagated into the column flange, giving the 

appearance of a through-thickness failure. Detailed fracture mechanics 

investigations conducted under this project confirm that damage initially 

identified as through thickness failures are likely to have occurred as a result of 

certain combinations of material strength notch toughness, conditions of stress in 

the connection, and the presence of critical flaws in the welded joint. 

As part of the research conducted in support of the development of these 

Recommended Criteria, extensive through-thickness testing of modern steels, 

meeting the ASTM A572, Gr. 50 and ASTM A913, Gr. 65 specifications has been 

conducted to determine the susceptibility of modern column materials to through 

thickness failures (FEMA-355A, State of the Art Report on Base Metals and 

Fracture). This combined analytical and laboratory research clearly shows that 

due to the restraint inherent in welded beam flange to column flange joints, the 

through thickness yield and tensile strengths of the column material is 

significantly elevated in the region of the connection. Further, for the modern 

materials tested, these strengths significantly exceed those that can be delivered 

to the column by beam material conforming to these same specifications. For this 

reason, no limits are suggested for the through thickness strength of modern steel 

materials. 

3.3.2.2 Base Material Toughness 

Material in rolled shapes with flanges 1-1/2 inches or thicker, and sections made from plates 

that are 2 inches or thicker, should be required to have minimum Charpy V-notch (CVN) 

toughness of 20 ft-lbs at 70� F. Refer to FEMA-353, Recommended Specifications and Quality 

Assurance Guidelines for Steel Moment-Frame Construction for Seismic Applications. 

Commentary: The 1997 AISC Seismic Provisions specify minimum notch 

toughness for rolled shapes with flanges 1-1/2 inches thick or thicker and sections 

made from plates 1-1/2 inches thick or thicker. These Recommended Criteria 

relax the requirement for toughness of plate material to apply to plates 2 inches 

or thicker as this was the original intent of the 1997 AISC specification, and it is 

believed that the AISC document will be revised to this requirement. 

Research has not clearly demonstrated the need for a specific value of base metal 

notch toughness. However, it is judged that base metal toughness is important to 

prevention of brittle fracture of the base metal in the highly stressed areas of the 

connection. A number of connection assemblies that have been tested have 

demonstrated base metal fractures at weld access holes and at other 

discontinuities such as at the ends of cover plates. In at least some of these tests, 

the fractures initiated in zones of low notch toughness. Tests have not been 

conducted to determine if higher base metal notch toughness would have reduced 

the incidence of such fractures. 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

The CVN value of 20 ft.-lbs at 70� F recommended here was chosen because it 

is usually achieved by modern steels, and because steels meeting this criterion 

have been used in connections which have performed successfully. Since current 

studies (FEMA-355A, State of the Art Report on Base Metals and Fracture) have 

indicated that rolled shapes produced from modern steels meet this requirement 

almost routinely, even in the thicker shapes that currently require testing, it has 

been suggested that the requirement for this testing could be eliminated and 

replaced by a certification program administered by the mills. However, such a 

program is not currently in existence. Until such time as such a certification 

program is in place, or a statistically meaningful sampling from all major mills 

has been evaluated, it is recommended that the AISC requirement for testing be 

continued. According to the Commentary to the 1997 AISC Seismic Provisions, 

thinner sections are judged not to require testing because they “are generally 

subjected to enough cross-sectional reduction during the rolling process that the 

resulting notch toughness will exceed that required.” In other words, the 

toughness is desired, but testing to verify it on a project basis is not judged to be 

necessary as it is routinely achieved. 

3.3.2.3 k-Area Properties 

The k-area of rolled wide-flange shapes, which may be considered to extend from the midpoint of the radius of the fillet from the flange into the web, approximately 1 to 1-1/2 inches 

beyond the point of tangency between the fillet and web, as defined in Figure C-6.1 of 1997 AISC 

Seismic Provisions, is likely to have low toughness and may therefore be prone to cracking 

caused by welding operations. Designers should detail welds of continuity plates and web 

doubler plates in columns in such a way as to avoid welding directly in the k-area. Refer to 

Section 3.3.3 for more information. 

Fabricators should exercise special care when making welds in, or near to, the k-area. Where 

welding in the k-area of columns cannot be avoided, special nondestructive testing is 

recommended. Refer to FEMA-353, Recommended Specifications and Quality Assurance 

Guidelines for Steel Moment-Frame Construction for Seismic Applications. 

Commentary: Recent studies, instigated in response to fabrication problems have 

shown that, for rotary-straightened W-shapes, an area of low material toughness 

can occur in the region of the web immediately adjacent to the flange. In some 

instances, cracking has occurred in these areas during welding. The Commentary 

to the 1997 AISC Seismic Provisions provides a figure (Fig. C-6.1) that defines 

the k-area. 

The low toughness of the k-area seems to be associated only with rotarystraightened sections. Which sections are rotary straightened varies among the 

mills. One major domestic supplier rotary-straightens all shapes weighing less 

than 150 pounds per linear foot. Larger sections are often straightened by other 

means that do not result in as much loss of toughness in the k-area. Because 

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rolling mill practice is frequently changed, it is prudent to assume that all rolled 

sections are rotary-straightened. 

3.3.2.4 Weld Metal Matching and Overmatching 

The use of weld metal with tensile strength that is significantly less than the expected 

strength of the connected base steel material is not recommended. Welding consumables 

specified for CJP groove welds of beam flanges and flange reinforcements should have yield and 

tensile strengths that are approximately the same as, or slightly higher than, the expected yield 

and tensile strength of the beam or girder flanges being welded. Significant overmatching of 

weld metal should not be required unless overmatching is specified in the connection 

prequalification or is used in the prototypes tested for project-specific qualification of the 

connection being used. Flux-Cored Arc Welding and Shielded Metal Arc Welding filler metals 

commonly used in structural construction and conforming to the E70 specifications provide 

adequate properties for joining most material conforming to ASTM A36, A572, Grades 42 and 

50, A913 Grade 50 and A992. Weld splices of columns conforming to ASTM A913, Grade 65 

steel should be made with filler metals capable of depositing weld metal with a minimum tensile 

strength of 80 ksi. 

Commentary: Undermatched weld metals, that is, weld metals with lower 

strength than the connected base metals, are beneficial in some applications in 

that they tend to limit the residual stress state in the completed joint. This can be 

achieved by employing balanced, or slightly undermatched filler metals. 

However, in applications where yield level stresses are anticipated, it is desirable 

to minimize the amount of plasticity in the welded joint. This can be achieved by 

employing balanced, or slightly overmatched filler metals. There is significant 

variation in the yield and tensile strengths of typical structural steels. Although 

E70 filler metals will produce matching or slightly overmatching conditions for 

most structural steel conforming to grade 36 and grade 50 specifications, they 

will not always provide these conditions. The new A992 specification for grade 

50 structural steel has controlled upper limits on the strength of the material and 

should produce, with E70 filler metals, more closely matching conditions in most 

cases. Notwithstanding the above, the majority of the successful connection tests 

performed under this project have used weld metals with yield and tensile 

strengths in the nominal range of 58 and 70 ksi respectively, and these have 

performed in an adequate manner. Therefore, it is believed that the use of E70 

filler metals with grade 50 structural steels is acceptable. For additional 

information, refer to FEMA-355B, State of the Art Report on Welding and 

Inspection. 

3.3.2.5 Weld Metal Toughness 

For structures in which the steel frame is normally enclosed and maintained at a temperature 

of 50o F or higher, critical welded joints in seismic-force-resisting systems, including CJP welds 

of beam flanges to column flanges, CJP groove welds of shear tabs and beam webs to column 

flanges, column splices, and similar joints, should be made with filler metal providing CVN 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

toughness of 20ft-lbs at -20� F and 40ft-lbs at 70� F and meeting the Supplemental Toughness 

Requirements for Welding Materials included in FEMA-353, Recommended Specifications and 

Quality Assurance Guidelines for Steel Moment-Frame Construction for Seismic Applications. 

For structures with lower service temperatures than 50o F, these qualification temperatures should 

be reduced accordingly. 

Commentary: Principles of fracture mechanics demonstrate the importance of 

notch toughness to resist fracture propagation from flaws, cracks, and backing 

bars or other stress concentrations that may be pre-existing or inherent, or that 

may be caused by applied or residual stresses. The 1997 AISC Seismic 

Provisions require the use of welding consumables with a rated Charpy V-Notch 

toughness of 20 ft.-lbs. at -20� F, for Complete Joint Penetration groove welds 

used in the seismic-force-resisting system. The 1997 AISC Seismic Provisions, 

Supplement No. 1, February 15, 1999 (AISC, 1999), changes this requirement to 

include “all welds used in primary members and connections in the SeismicForce-Resisting System”. The rating of the filler metal is as determined by AWS 

classification or manufacturer certification. 

Studies conducted under this project have indicated that not all weld 

consumables that are rated for 20 ft-lbs of toughness at –20oF will provide 

adequate toughness at anticipated service temperatures. The supplemental 

toughness requirements contained in FEMA-353 are recommended to ensure that 

weld metal of adequate toughness is obtained in critical joints. 

Most of the beam-column connection tests conducted under this project were 

made with filler metal conforming to either the E70T6 or E70TGK2 designations. 

These filler metals generally conform to the recommended toughness 

requirements. Other filler metals may also comply. 

3.3.2.6 Weld Backing, Weld Tabs, and Other Welding Details 

Weld backing and runoff tabs should be removed from CJP flange welds, unless otherwise 

noted in the connection prequalification or demonstrated as not required by project-specific 

qualification testing. Refer to FEMA-353, Recommended Specifications and Quality Assurance 

Guidelines for Steel Moment-Frame Construction for Seismic Applications, for special 

requirements for weld backing, weld tabs and other welding details for moment-frame joints. 

Commentary: It was originally hypothesized, following the 1994 Northridge 

earthquake that weld backing created an effective crack equal to the thickness of 

the backing and that this phenomena was responsible for many of the fractures 

that had occurred. Finite element analyses of welded joints (Chi, et. al., 1997) 

have shown that although the backing does create some notch effect, far more 

significant is the fact that when backing is left in place, it obscures effective 

detection of significant flaws that may exist at the weld root. These flaws 

represent a significantly more severe notch condition than does the backing itself. 

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It is recommended that backing be removed from beam bottom-flange joints, 

to allow identification and correction of weld root flaws. This is not required for 

top-flange joints because the stress condition at the root of the top flange weld is 

less critical and less likely to result in initiation of fracture, even if some weld 

root flaws are present. Also, as a result of the more favorable position, it is far 

less likely that significant flaws will be incorporated in top-flange joints. Special 

welding of backing for top-flange welds is recommended. 

Weld tabs represent another source of potential discontinuity at the critical 

weld location. Additionally, the weld within the weld tab length is likely to be of 

lower quality and more prone to flaws than the body of the weld. Flaws in the 

runoff tab area can create stress concentrations and crack starters and for this 

reason their removal is recommended. It is important that the process of removal 

of the weld tabs not be, of itself, a cause of further stress concentrations, and 

therefore, FEMA-353 requires that the workmanship result in smooth surfaces, 

free of defects. 

3.3.2.7 Weld Access Holes 

New welded moment-resisting connections should utilize weld access hole configurations as 

shown in Figure 3-5, except where otherwise noted in specific connection prequalifications. 

Criteria for forming and finishing of weld access holes are provided in FEMA-353, 

Recommended Specifications and Quality Assurance Guidelines for Steel Moment-Frame 

Construction for Seismic Applications. 

Commentary: The size, shape and workmanship of weld access holes can affect 

the connection in several different ways. If the hole is not large enough, this 

restricts welder access to the joint and increases the probability of low quality 

joints. Depending on the size and shape of the weld access hole, plastic strain 

demands in the welded joint and in the beam flange at the toe of the weld access 

hole can be significantly affected. Laboratory tests of unreinforced connections 

fabricated with tough filler metals have indicated that these connections 

frequently fail as a result of low cycle fatigue of the beam flange material at the 

toe of the weld access hole, resulting from the strain concentrations introduced by 

this feature. The configuration shown in Figure 3-5 was developed as part of a 

program of research conducted in support of the development of these 

Recommended Criteria and appears to provide a good balance between adequate 

welder access and minimization of stress and strain concentration. For further 

discussion of weld access holes, see FEMA-355D, State of the Art Report on 

Connection Performance. 

3.3.2.8 Welding Quality Control and Quality Assurance 

FEMA-353, Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications, includes recommendations for quality 

control and quality assurance for steel moment frames and connections intended for seismic 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

applications. Recommended inspections are divided into two categories: Process and Visual 

Inspection, and Nondestructive Testing. For each category, different levels of quality assurance 

measures are specified depending on the anticipated severity of loading, or demand (Seismic 

Weld Demand Category) and the consequence of welded joint failure (Seismic Weld 

Consequence Category). All welded joints in the seismic-force-resisting system should be 

categorized according to the applicable Consequence and Demand Categories, using the 

following form: “QC/QA Category – BH/T”, where the first letter (in this case B) indicates the 

Demand Category, the second letter (in this case H) indicates the Consequence Category and the 

third letter, either T or L, indicates that primary loading is either transverse or longitudinal, 

respectively. The various categories are described in detail in the referenced documents. For the 

prequalified connections included in these Recommended Criteria, the appropriate categories 

have been preselected and are designated in the information accompanying the prequalfication. 

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Criteria for New Steel FEMA-350 

olerances shall not 

ccumulate to the extent 

at the angle of the access 

ole cut to the flange 

urface exceeds 25�. 

Notes: 

1. Bevel as required by AWS D1.1 for selected groove weld procedure. 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

2. Larger of tbf or 1⁄2 inch. (plus 1⁄2 tbf , or minus 1⁄4 tbf) 

3. 3⁄4 tbf to tbf, 3⁄4” minimum (– 1⁄4 inch) 

4. 3/8” minimum radius (plus not limited, or minus 0) 

5. 3 tbf . (– 1⁄2 inch) 

6. See FEMA-353, Recommended Specifications and Quality Assurance Guidelines for Steel Moment-Frame 

Construction for Seismic Applications, for fabrication details including cutting methods and smoothness 

requirements. 

Figure 3-5 Recommended Weld Access Hole Detail 

Commentary: FEMA-353 describes the Demand (A,B,C) and Consequence 

(H,M,L) Categories and indicates the appropriate levels of Visual and NDT 

inspection for each combination of demand and consequence category. The 

degree of inspection recommended is highest for the combination of high demand 

(Category A) with high consequence (Category H) welded joints, and conversely, 

less inspection is required for low demand (Category C) with low consequence 

(Category L) welded joints, with intermediate levels for categories in between. 

3.3.3 Other Design Issues for Welded Connections 

3.3.3.1 Continuity Plates 

Unless project-specific connection qualification testing is performed to demonstrate that 

beam flange continuity plates are not required, moment-resisting connections should be provided 

with beam flange continuity plates across the column web when the thickness of the column 

flange is less than the value given either by Equation 3-5 or 3-6: 

tcf < 0.4 18 yb yb 

ff 

yc yc 

FR 

.bt 

FR (3-5)

b 

t < (3-6) f 

cf 6

where: 

tcf = minimum required thickness of column flange when no continuity plates 

are provided, inches 

bf = beam flange width, inches 

tf = beam flange thickness, inches 

Fyb (Fyc) = Minimum specified yield stress of the beam (column) flange, ksi 

Ryb (Ryc) = the ratio of the expected yield strength of the beam (column) material to the 

minimum specified yield strength, as in the 1997 AISC Seismic Provisions. 

Where continuity plates are required, the thickness of the plates should be determined 

according to the following: 

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• For one-sided (exterior) connections, continuity plate thickness should be at least one-half of 

the thickness of the two beam flanges. 

• For two-sided (interior) connections, the continuity plates should be equal in thickness to the 

thicker of the two beam flanges on either side of the column. 

• The plates should also conform to Section K1.9 of AISC-LRFD Specifications. 

Continuity Plates should be welded to column flanges using CJP groove welds as shown in 

Figure 3-6. Continuity plates should be welded to the web as required to transmit the shear 

capacity of the net length of the continuity plate. 

Notes 

1. Web doubler plate where required by Section 3.3.3.2. See 1997 AISC Seismic Provisions Section 9.3c, 

Commentary C9.3, and Figures C-9.2 and C-9.3 for options and connection requirements. Weld QC/QA 

Category BL/L for all welds. 

2. Continuity plate as required by Section 3.3.3.1. 

3. Required total weld strength = ( ) pl pl Lnet Fy 0 .6t . QC/QA Category BL/L. 

4. CJP typical. QC/QA Category BM/T. For exterior beam-column connections (beam one side only), weld 

of continuity plate to column flange at free side may be fillet welds at top and bottom face of plate. 

5. AISC minimum continuous fillet weld under backing. 

6. Minimum width to match beam flange. Preferred alternative: extend plate flush with column flanges. 

7. Remove weld tabs to 1⁄4” maximum from edge of continuity plate. Grind end of weld smooth (500 m-in), 

not flush. Do not gouge column flange. 

8. Beam connection, see Figures 3-7 through 3-20. 

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Figure 3-6 Typical Continuity and Doubler Plates 

Commentary: Following the 1994 Northridge earthquake, some engineers 

postulated that the lack of continuity plates was a significant contributing factor 

to the failure of some connections. This was partially confirmed by initial tests 

conducted in 1994 in which several specimens without continuity plates failed 

while some connections with these plates successfully developed significant 

ductility. Based on this, FEMA-267 recommended that all connections be 

provided with continuity plates. The 1997 AISC Seismic Provisions, published 

after FEMA-267, relaxed this criteria and state that continuity plates should be 

provided to match those in connections tested to obtain qualification. 

Research conducted in this project suggests that where the flange thickness of 

columns is sufficiently thick, continuity plates may not be necessary. Equation 3-5 

was the formula used by AISC to evaluate column flange continuity plate 

requirements prior to the Northridge earthquake. It appears that this formula is 

adequate to control excessive column flange prying as long as the beam flanges 

are not too wide. Studies reported in FEMA-355D suggest that the ratio of beamflange width to column-flange thickness is also important. Tests with a ratio of 

5.3 (W36x150 beam with W14x311 column) showed little difference in 

performance with or without continuity plates, while tests with a ratio of 6.8 

(W36x150 beam with W27x258 column) showed some difference of performance. 

The factor of 6 in Equation 3-6 was selected, based on these tests and engineering 

judgement. 

3.3.3.2 Panel Zone Strength 

Moment-resisting connections should be proportioned either so that shear yielding of the 

panel zone initiates at the same time as flexural yielding of the beam elements or so that all 

yielding occurs in the beam. The following procedure is recommended: 

Step 1: Calculate t, the thickness of the panel zone that results in simultaneous yielding of 

the panel zone and beam from the following relationship: 

h d - CM b 

y c 

t =

(0.9) 0.6F Rycd

h

c ( db - t fb ) (3-7) yc 

where: 

h = the average story height of the stories above and below the panel zone. 

Ryc = the ratio of the expected yield strength of the column material to the minimum 

specified yield strength, in accordance with the 1997 AISC Seismic Provisions. 

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Mc and Cy are the coefficients defined in Section 3.2.6 and Section 3.2.7, respectively, and 

other terms are as defined in AISC-LRFD 

Step 2: If t, as calculated, is greater than the thickness of the column web, provide doubler 

plates, or increase the column size to a section with adequate web thickness. 

Where doubler plates are required, the thickness should be determined as described above, 

and they should be proportioned and welded as described in the 1997 AISC Seismic Provisions. 

QC/QA Category BL/L procedures are defined in FEMA-353. For connections designed using 

project-specific qualifications, the panel zone strength should match that of the tested 

connections. 

Commentary: Several aspects of the methodology for the design of panel zones, 

as contained in the 1997 AISC Seismic Provisions, are considered to require 

revision, based on studies conducted by this project. As described in FEMA355D, the best performance is likely to be achieved when there is a balance of 

beam bending and panel zone distortion. The equations given are intended to 

provide panel zones that are just at the onset of yielding at the time the beam 

flange begins to yield. 

The procedure recommended in these Recommended Criteria differs 

significantly from that contained in the 1997 AISC Seismic Provisions, but the 

results are not dramatically different. For most column sizes results will be 

similar to methods used in the past. For columns with thick flanges, the methods 

herein will result in the need for moderately thicker panel zones than in the past. 

3.3.3.3 Connections to Column Minor Axis 

Connections to the minor axis of a column should be qualified by testing following the 

procedures of Section 3.9. If minor-axis connections are to be used in conjunction with majoraxis connections at the same column, the testing program should include biaxial bending effects 

at the connection. 

Commentary: In general, the prequalified connections have not been tested for 

use with columns oriented so that beams connect to the minor axis of the column. 

Two tests of Reduced Beam Section connections in this orientation were 

conducted, which indicated good performance. These tests were conducted to 

provide a general indication of the possible performance of weak axis 

connections, but are not considered to comprise a sufficient database for 

prequalification of such connections. 

3.3.3.4 Attachment of Other Construction 

Welded or bolted attachment for exterior facades, partitions, ductwork, piping, or other 

construction should not be placed in the hinging area of moment frame beams. The hinging area 

is defined as one half of the beam depth on either side of the theoretical hinge point as described 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

in the prequalification data table for each connection detail. It is recommended that bolt holes for 

this type of construction not be permitted between the face of the column and six inches, 

minimum, beyond the extreme end of the hinging area. Outside of the described area, a 

calculation should be made to assure sufficient net section to avoid fracture, based on moments 

calculated using the expected moment at the hinge point. Welding between the column face and 

the near edge of the hinging area should be carefully controlled to avoid creation of stress 

concentrations and application of excessive heat. Specifications and drawings should clearly 

indicate that anchorage shall not be made in the areas described and this should be coordinated 

with the architect and other members of the design team. 

Commentary: It is common for precast panels and other facade elements, as well 

as other construction, to be anchored to members of the steel frame through the 

use of welds, bolts, powder-driven fasteners, or other fasteners. Such anchorage 

is often not considered by the engineer and is not performed with the same care 

and quality control as afforded the main building structure. Such anchorage, 

made in an area of high stress, can create stress concentrations leading to 

potential fracture. 

3.3.4 Bolted Joints 

The 1997 AISC Seismic Provisions contain requirements for bolted joints used in seismicforce-resisting systems. These requirements should be followed, as supplemented by the specific 

requirements given in the individual design procedures provided for the prequalified bolted 

connections, or where special bolting requirements are used in project-specific tested 

connections. QA/QC requirements for bolted joints are given in FEMA-353, Recommended 

Specifications and Quality Assurance Guidelines for Steel Moment-Frame Construction for 

Seismic Applications. Where these Recommended Criteria require, or permit, the use of bolts 

conforming to ASTM A325, an acceptable alternative is twistoff bolts conforming to ASTM 

F1852. 

3.4 Prequalified Connections – General 

Prequalified connection details are permitted to be used for moment frame connections for 

the types of moment frames and ranges of the various design parameters indicated in the limits 

accompanying each prequalification. Project-specific testing should be performed to 

demonstrate the adequacy of connection details that are not listed herein as prequalified, or are 

used outside the range of parameters indicated in the prequalification. Designers should follow 

the procedures outlined in Section 3.9 for qualification of nonprequalified connections. 

Commentary: The following criteria were applied to connections listed as 

prequalified: 

1. There is sufficient experimental and analytical data on the connection 

performance to establish the likely yield mechanisms and failure modes for 

the connection. 

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Moment-Frame Buildings Chapter 3: Connection Qualification

Criteria for New Steel FEMA-350 

2. Rational models for predicting the resistance associated with each 

mechanism and failure mode have been developed. 

3. Given the material properties and geometry of the connection, a rational 

procedure can be used to estimate which mode and mechanism controls the 

behavior and the deformation capacity (that is, interstory drift angle) that can 

be attained from the controlling conditions. 

4. Given the models and procedures, the existing data base is adequate to 

permit assessment of the statistical reliability of the connection. 

Some of the connections in the following sections are only prequalified for use in 

Ordinary Moment Frames, while others are prequalified for both Ordinary 

Moment Frame (OMF) and Special Moment Frame (SMF) use. In general, when 

a connection is qualified for use in SMF systems, it is also qualified for use in 

OMF systems with fewer restrictions on size, span, and other parameters than 

are applied to the SMF usage. For SMF application, very little extrapolation 

beyond the parametric values for which testing has been performed has been 

applied. For OMF application, some judgement has been applied to permit 

extrapolation for OMFs, based on the significantly lower rotational demands 

applicable to those systems. 

Some connection types for which extensive testing has been performed have not 

been included as prequalified for new buildings. These include the following: 

1. Welded Cover Plated Flange (WCPF); 

2. Welded Bottom Haunch (WBH); 

3. Welded Top and Bottom Haunch (WTBH). 

In general, these connections are not included because they do not have any 

significant advantages in performance over connections that are much simpler 

and cost-effective. The haunched connections in particular were not studied in 

detail by this project, because they were not considered to be economically 

practical for application in new buildings. Consequently, the data base for this 

connection type is insufficient to permit prequalification. WCPF connections, 

similarly, are relatively expensive, and, although there is a fairly large data base 

of tests, many of them successful, there have also been some significant brittle 

failures of this type of connection. The fact that these connections are not listed 

in this guideline as prequalified is not intended to preclude their use, nor to 

suggest that those structures for which they have been used previously are not 

expected to exhibit acceptable performance. Rather, it is believed that for new 

construction there are connections which are equally or more reliable, yet less 

costly. For those desiring to use the above listed connections, information is 

provided in FEMA-355D, State of the Art Report on Connection Performance, 

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FEMA-350 Criteria for New Steel 

and in the specific laboratory test reports referenced therein. Design procedures 

are given in FEMA-351 for some of these connections for use in upgrading 

existing buildings. 

Prequalified connections are also recommended for use without further 

testing in structures having dual systems, as defined in FEMA-302, provided that 

attachment of bracing to the connection does not inhibit or alter the yield 

mechanism for the assembly. 

3.4.1 Load Combinations and Resistance Factors 

Design procedures for prequalified connection upgrades contained in Sections 3.5, 3.6 and 

3.7 of this document are formatted on an expected strength basis, as opposed to either a Load and 

Resistance Factor Design basis or Allowable Stress Design basis. Loading used in these design 

formulations is generally calculated on the basis of the stresses induced in the assembly at 

anticipated yielding of the beam-column connection assembly. Where these design procedures 

require that earthquake loading be applied simultaneously with dead and live loading, the 

applicable load combinations of the 1997 AISC Seismic Provisions apply. Resistance factors 

should not be applied except as specifically required by the individual design procedure. 

3.5 Prequalified Welded Fully Restrained Connections 

This section provides prequalification data and design procedures for alternative types of 

welded, fully restrained, steel moment-frame connections, suitable for use in new construction. 

Table 3-1 indicates the various types of prequalified fully restrained connections, and the 

structural systems for which they are prequalified. Additional prequalification data on these 

connections are provided in the following sections. 

Table 3-1 - Prequalified Welded Fully Restrained Connections 

Connection Type Criteria Section Frame Type 

Welded Unreinforced Flanges – Bolted Web (WUF-B) 3.5.1 OMF 

Welded Unreinforced Flanges – Welded Web (WUF-W) 3.5.2 OMF, SMF 

Free Flange (FF) 3.5.3 OMF, SMF 

Reduced Beam Section (RBS) 3.5.4 OMF, SMF 

Welded Flange Plate (WFP) 3.5.5 OMF, SMF 

Commentary: FEMA-355D, State of the Art Report on Connection Performance, 

provides extensive information on the testing and performance of these 

connections, as well as others, that is not repeated in this document. The data 

presented in FEMA-355D have been prepared in support of the development of 

prequalification performance data, design procedures and limitations on design 

parameters for these connections. The design recommendations contained in 

FEMA-355D will not in all cases be identical to those contained herein. In some 

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Moment-Frame Buildings Chapter 3: Connection Qualification

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cases, the format, notation, and context of the design formulae contained in 

FEMA-355D have been modified to provide for consistent application within the 

design procedures of these Recommended Criteria. 

3.5.1 Welded Unreinforced Flange – Bolted Web Connections 

This section provides recommended criteria for design of fully restrained, Welded 

Unreinforced Flange – Bolted Web (WUF-B) connections. This type of connection is 

prequalified only for Ordinary Moment Frame applications, and within the parameters given in 

Table 3-2. 

WUF-B connections utilize complete joint penetration (CJP) groove welds, meeting the 

requirements of FEMA-353, Recommended Specifications and Quality Assurance Guidelines for 

Steel Moment Frame Construction for Seismic Applications, to join beam or girder flanges 

directly to column flanges. In this type of connection, no element other than weld metal, is used 

to join the flanges. Weld access holes are configured as indicated in Section 3.3.2.7. Web joints 

for these connections are made with slip-critical, high-strength bolts connecting the beam web to 

a shear tab that is welded to the column flange. Figure 3-7 provides a typical detail for this 

connection type. These connections should be designed in accordance with the criteria of this 

section. 

Notes 

1. See Figure 3-8 and Note 1 of Figure 3-8 for top and bottom flange weld requirements. QC/QA category 

AH/T. Refer to Figure 3-5 for weld access hole detail. 

2. Bolted shear tab. Use pretensioned A325 or A490 bolts. Weld to column flange with fillet weld both 

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sides, or with CJP weld, to develop full shear strength of plate. Weld QC/QA Category BL/T. 

3. See Figure 3-6 for continuity plate and web doubler plate requirements. 

Figure 3-7 Welded Unreinforced Flange – Bolted Web (WUF-B) Connection 

Commentary: This connection closely resembles the “prescriptive connection” 

commonly in use prior to the 1994 Northridge earthquake. After significant 

study, it has been concluded that with several improvements and appropriate 

levels of quality assurance with regard to workmanship and materials, this 

connection can perform reliably in frames designed as Ordinary Moment Frames 

(OMF) within the limitations indicated in Table 3-2. 

The improvements incorporated in this connection over typical connections 

detailed prior to the 1994 Northridge earthquake include the following: 

1. Weld metal with appropriate toughness; 

2. Removal of weld backing from bottom-beam-flange-to-column-flange welds, 

back-gouging and addition of a reinforcing fillet weld; 

3. Use of improved weld access hole shape and finish; 

4. Improvements to weld quality control, and quality assurance requirements 

and methods. 

Table 3-2 Prequalification Data WUF-B Connections 

General: 

Applicable systems Ordinary Moment Frame 

Hinge location distance sh dc /2 + db/2 

Critical Beam Parameters: 

Maximum depth W36 and shallower 

Minimum span-to-depth ratio 7 

Flange thickness 1” maximum 

Permissible material specifications A572 Grade 50, A992, A913 Grade 50/S75 

Critical Column Parameters: 

Depth W8, W10, W12, W14 

Permissible material specifications A572 Grade 50; A913 Grade 50 and 65; A992 

Beam/Column Relations: 

Panel Zone strength Section 3.3.3.2 

Column/beam bending strength No Requirement (OMF) 

Connection Details 

Web connection Shear tab welded to column, bolted to beam. 

Continuity plate thickness Section 3.3.3.1 

Flange welds See Fig. 3-8 and Section 3.3.2.5 

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Welding parameters Section 3.3.2.4, 3.3.2.5, 3.3.2.6 

Weld access holes Section. 3.3.2.7 

For best performance of this connection some limited panel zone yielding is 

beneficial. For this reason, it is recommended that panel zones not be overreinforced. 

3.5.1.1 Design Procedure 

Step 1: Calculate Mpr, at hinge location, sh, according to methods of Section 3.2.4. 

Step 2: Calculate Vp, at hinge location, sh, according to methods of Section 3.2.5. 

Step 3: Calculate Mc, Mf, and Cy as described in Section 3.2.6 and 3.2.7. 

Step 4: Calculate the required panel zone thickness using the procedures of Section 3.3.3.2. 

Step 5: Calculate the connection shear as : 

2M 

Vf = 

f + V (3-8) L - d g 

c 

where: 

Vf = maximum shear at the column face, kips 

Vg = shear at the column face due to factored gravity loads, kips. 

Step 6: Design the shear tab and bolts for Vf. Bolts should be designed for bearing, using a 

resistance factor f of unity 

Step 7: Check requirements for continuity plates according to Section 3.3.3.1. 

Step 8: Detail the connection as shown in Figure 3-7 and Note 1 of Figure 3-8. 

3.5.2 Welded Unreinforced Flange – Welded Web Connections 

This section provides guidelines for design of fully restrained, Welded Unreinforced Flange – 

Welded Web (WUF-W) connections. This type of connection is prequalified for use in Ordinary 

Moment Frame and Special Moment Frame systems within the parameters given in Table 3-3. 

These connections utilize complete joint penetration (CJP) groove welds, meeting the 

requirements of FEMA-353, Recommended Specifications and Quality Assurance Guidelines for 

Steel Moment-Frame Construction for Seismic Applications, to join beam flanges or girder 

flanges directly to column flanges. In this type of connection, no reinforcement is provided 

except for the addition of a fillet weld applied to the groove weld. Web joints for these 

connections are made with complete joint penetration groove welds of the beam web to the 

column flange. Weld access holes for this type of connection should be in accordance with 

Section 3.3.2.7. Figure 3-8 provides a typical detail for this connection type. These connections 

should be designed in accordance with the procedures of this section. 

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Notes 

1. CJP groove weld at top and bottom flanges. At top flange, either (1) remove weld backing, backgouge, 

and add 5/16” minimum fillet weld, or (2) leave backing in place and add 5/16” fillet under backing. At 

bottom flange, remove weld backing, backgouge, and add 5/16” minimum fillet weld. Weld: QC/QA 

Category AH/T. 

2. Weld access hole, see Figure 3-5. 

3. CJP groove weld full length of web between weld access holes. Provide non-fusible weld tabs. Remove 

weld tabs after welding and grind end of weld smooth at weld access hole. Weld: QC/QA Category 

BH/T. 

4. Shear tab of thickness equal to that of beam web. Shear tab length shall be so as to allow 1/8” overlap 

with the weld access hole at top and bottom, and the width shall extend 2” minimum back along the 

beam, beyond the end of the weld access hole. 

5. Full-depth partial penetration from far side. Weld: QC/QA Category BM/T. 

6. Fillet weld shear tab to beam web. Weld size shall be equal to the thickness of the shear tab minus 1/16”. 

Weld shall extend over the top and bottom one-third of the shear tab height and across the top and 

bottom. Weld: QC/QA Category BL/L. 

7. Erection bolts: number, type, and size selected for erection loads. 

8. For continuity plates and web doubler plates see Figure 3-6. 

Figure 3-8 Welded Unreinforced Flange-Welded Web (WUF-W) Connection 

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Table 3-3 Prequalification Data WUF-W Connections 

General: 

Applicable systems OMF, SMF 

Hinge location distance sh dc /2 + db/2 

Critical Beam Parameters: 

Maximum depth W36 and shallower 

Minimum span-to-depth ratio OMF: 5 

SMF: 7 

Flange thickness OMF: 1-1/2”or less 

SMF: 1” or less 

Permissible material specifications A572 Grade 50, A992, A913 Grade 50/S75 

Critical Column Parameters: 

Depth OMF: Not Limited 

SMF: W12, W14 

Permissible material specifications A572 Grade 50; A913 Grade 50 and 65; A992 

Beam/Column Relations: 

Panel Zone strength SMF: Section 3.3.3.2 

Column/beam bending strength Section 2.9.1 

Connection Details 

Web connection Special Connection – See Fig. 3-8 

Continuity plate thickness Section 3.3.3.1 

Flange welds Section 3.3.2.5 

Welding parameters Section 3.3.2.4, 3.3.2.5, 3.3.2.6 

Weld access holes Section. 3.3.2.7 

Commentary: Development of connections with unreinforced flanges, suitable for 

use in Special Moment Frames, has required significant research, resulting in 

major modifications to the connection commonly in use prior to the 1994 

Northridge earthquake. A summary list of revisions to the original prescriptive 

connection incorporated in this detail is as follows: 

1. limitations on permitted beam sizes, 

2. filler metal with appropriate toughness, 

3. removal of weld backing, back-gouging and addition of a reinforcing fillet 

weld, 

4. use of improved weld-access hole shape and finish, 

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5. improvements to weld quality control and quality assurance requirements and 

methods, and 

6. use of a full-strength welded web joint. 

Research indicates that this type of connection can be constructed to perform 

reliably if all of the procedures are complied with. Although this connection may 

appear to be economical, compared with other prequalified details, the designer 

should note carefully the importance of the features of this detail that improve its 

performance, and consider the effects of these features on the connection cost, 

before selecting it as a standard. Of particular importance is the rigorous level of 

quality assurance during field erection and welding, required for successful 

performance of this connection. Additionally, the beam size limitations may make 

it impractical in some buildings. 

3.5.2.1 Design Procedure 

Step 1: Calculate Mpr, at hinge location, Sh, according to methods of Section 3.2.4.

Step 2: Calculate Vp, at hinge location, Sh, according to methods of Section 3.2.5.

Step 3: Calculate Mc and Cy as described in Sections 3.2.6 and 3.2.7, respectively.

Step 4: Calculate the required panel zone thickness using the procedures of Section 3.3.3.2.

Step 5: Check requirements for continuity plates according to Section 3.3.3.1.

Step 6: Detail the connection as shown in Figure 3-8.

3.5.3 Free Flange Connections 

This section provides guidelines for design of fully restrained Free Flange (FF) connections. 

This type of connection is prequalified for use in Special Moment Frame systems for beam sizes 

within the limits given in Table 3-4. For larger beams, the connection is prequalified for use in 

Ordinary Moment Frame systems. 

These connections utilize complete joint penetration groove welds, meeting the requirements 

of FEMA-353, Recommended Specifications and Quality Assurance Guidelines for Steel Moment 

Frame Construction for Seismic Applications, to join beam flanges or girder flanges directly to 

column flanges. The web of the beam is removed in a single cut in the area adjacent to the 

column flange, and is replaced with a heavy trapezoidal-shaped shear tab. The shear tab is CJP 

groove welded to the column flange and welded on all sides with a fillet weld to the beam web. 

Figure 3-9 provides a typical detail for this connection type. These connections should be 

designed in accordance with the guidelines of this section. 

Commentary: This connection type was developed at the University of Michigan 

and has been extensively tested both at that university and at the University of 

Texas at Austin. This connection type has demonstrated good performance, 

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Criteria for New Steel FEMA-350 

similar to that exhibited by the WUF-W connection described in Section 3.5.2, 

and, in fact, has many similarities to that connection, as follows: 

1. The flange weld is the same as the WUF-W; 

2. The web cut-out provides an improvement similar to that provided by the 

improved weld-access hole; 

3. The web connection is very substantial. 

Table 3-4 Prequalification Data for Free Flange Connections 

General: 

Applicable systems OMF, SMF 

Hinge location distance sh (dc+db)/2 

Critical Beam Parameters: 

Maximum depth OMF: W36 and shallower 

SMF: W30 and shallower 

Minimum span-to-depth ratio OMF: 5 

SMF: 7 

bf /2tf of flange 52/�Fy 

Flange thickness OMF: 1-1/4” and less 

SMF: 3⁄4” and less 

Permissible material specifications A572 Grade 50, A992, A913 Grade 50/S75 

Critical Column Parameters: 

Depth OMF: Not limited 

SMF: W12, W14 

Permissible material specifications A572 Grade 50; A913 Grade 50 and 65, A992 

Beam/Column Relations: 

Panel zone strength SMF: Section 3.3.3.2 ; Cpr=1.2 

Column/beam bending strength SMF: Section 2.9.1; Cpr=1.2 

Connection Details 

Web connection Heavy welded shear tab: See Figure 3-9 

Continuity plate thickness Section 3.3.3.1 

Flange welds Fig. 3-9 

Welding parameters Section 3.3.2.4, 3.3.2.5, 3.3.2.6 

Weld access holes Not applicable 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

Notes 

1. CJP groove weld. Note 1 of Figure 3-8 applies. Weld: QC/QA Category AH/T. 

2. See design procedure in Section 3.5.3.1, Steps 5 through 8, for web plate size and thickness. 

3. 1⁄2” minimum radius. 

4. Erection bolts: number, type and size selected for erection loads. 

5. CJP double-bevel groove weld. Weld: QC/QA Category BH/T. 

6. Fillet welds size, length, calculated in Section 3.5.3.1, Step 8. Weld: QC/QA Category BH/L. 

7. For continuity plates and web doubler plates see Figure 3-6. 

Figure 3-9 Welded Free Flange (FF) Connection 

3.5.3.1 Design Procedure 

Step 1: Calculate Mpr at hinge location, Sh, according to the methods of Section 3.2.4.

Step 2: Calculate Vp at hinge location, Sh, according to the methods of Section 3.2.5.

Step 3: Calculate Mf, Mc, and Cy as described in Sections 3.2.5, 3.2.6, and 3.2.7.

Step 4: Calculate the length of the free flange:

Lff = at fb (3-9) 

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where a may be selected in the range of 5 to 6. 

Step 5: Calculate the shear in the shear tab from the equation: 

2 M 

Vst = f +V (3-10) L -d g 

c 

where: 

Vst = shear in the shear tab, kips 

L = span length measured from center to center of columns, ft 

Vg = shear at the beam end due to factored gravity loads, kips 

Step 6: Calculate the tension force on the shear tab, Tst from the equation: 

M M f Tst = f -Tf = 

db -t fb 

-Ry Fyb bfb tfb (3-11) db -t fb 

Step 7: Calculate the required height of the shear tab from the equation: 

hst = db -2t fb -2b (3-12) 

where b = 2 inches 

Step 8: Calculate the required thickness of the shear tab and the weld sizes for the forces 

shown in Figure 3-10, based on principles of mechanics. Note that it is assumed that 

only the regions at the ends of the plate, and having a dimension db/4 are effective in 

resisting these forces. 

Step 9: Determine the required panel zone thickness according to the methods of Section 

3.3.3.2. 

Step 10: Check requirements for Continuity Plates according to Section 3.3.3.1. 

Step 11: Detail the connection as shown in Figure 3-9. 

3.5.4 Welded Flange Plate Connections 

This section provides guidelines for design of fully restrained Welded Flange Plate (WFP) 

connections. These connections utilize plates to connect the beam flanges to the column flange, 

without any direct connection of the beam flange to the column flange. The flange-plate-tocolumn-flange joint is a complete joint penetration groove weld. The flange plates are fillet 

welded to the top and bottom of the beam top and bottom flanges, respectively. Figure 3-11 

provides a typical detail for this type of connection. These connections should be designed in 

accordance with the procedures of this section. 

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Tst 

0.5Vst 

0.5Vst 

Tst 

Figure 3-10 Schematic of the Forces for Design of the Free Flange Shear Tab 

Table 3-5 Prequalification Data for WFP Connections 

General 

Applicable systems OMF, SMF 

Hinge location distance sh dc/2 + lp 

Critical Beam Parameters 

Maximum depth W36 and shallower 

Minimum span-to-depth ratio OMF: 5 

SMF: 7 

Flange thickness OMF: 1-1/2”or less 

SMF: 1”and less 

Permissible material specifications A572 Grade 50, A992, A913 Grade 50/S75 

Critical Column Parameters 

Depth OMF: Not limited 

SMF: W12, W14 

Permissible material specifications A572 Grade 50; A913 Grade 50 or 65, A992 

Beam/Column/Flange Plate (FP) Relations 

Panel Zone strength Section 3.3.3.2 

Column/beam bending strength ratio, Section 2.9.1 

Connection Details 

Flange plate size Section 3.5.4.1 

Flange plate material Grade 50 

Flange welding Fig. 3-11 

Flange plate filler metals Section 3.3.2.4 

Web connection Section 3.5.4.3 and Figure 3-11 

Web welding parameters Section 3.3.2.4 

Continuity plate thickness Sec 3.3.3.1, Consider dimensions of beam flange to 

be equal to dimension of flange plate. 

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Notes 

1. Flange plate. See Section 3.5.4.1, Steps 1-4, for sizing requirements. Plates shall be fabricated with 

rolling direction parallel to the beam. 

2. CJP groove weld: single or double bevel. Weld in shop or field. When using single-bevel groove weld, 

remove backing after welding, back-gouge, and reinforce with 5/16”-minimum fillet weld. When using 

double bevel weld, back-gouge first weld before welding other side. Weld QC/QA Category AH/T. If 

plates are shop welded to column, care must be exercised in locating and leveling plates, as shimming is 

not allowed between the plates and the beam flanges. If plates are field-welded to column after 

connecting to beam, weld access holes of sufficient size for weld backing and welding access shall be 

provided. 

3. Fillet welds at edges of beam flanges to plate. Size welds according to the procedure in Section 3.5.4.1, 

Step 5. Welds may be shop or field. Provide weld tabs at end to provide full weld throat thickness to the 

end of the plate. Remove weld tabs and grind the end of the weld smooth. Use care to avoid grinding 

marks on the beam flange. Weld: QC/QA Category BH/L. 

4. Fillet weld at end of flange plate to beam flange. Welds may be shop or field. Maintain full weld throat 

thickness to within 1” of the edge of the flange. Weld: QC/QA Category BH/T. 

5. Shear tab of length equal to db-2k–2”. Shear tab thickness should match that of beam web. 

6. Erection bolts: number, type, and size selected for erection loads. 

7. Full depth-partial penetration from far side. Weld: QC/QA Category BM/T. 

8. Fillet weld both sides. Fillet on side away from beam web shall be same size as thickness of shear tab. 

Fillet on the side of the beam web shall be 1⁄4”. Weld: QC/QA Category BH/T. 

9. Fillet weld shear tab to beam web. Weld size shall be equal to the thickness of the shear tab minus 1/16”. 

Weld: QC/QA Category BH/L. 

10. For continuity plates and web doubler plates see Figure 3-6. For calculation of continuity plate requirements, 

use flange plate properties instead of beam flange properties. 

Figure 3-11 Welded Flange Plate (WFP) Connection 

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Commentary: The WFP connection was tested at the University of California at 

Berkeley. Several similar connections had been tested by private parties prior to 

testing under this project. The connection has similarities to both the cover 

plated connection, which has been extensively used, and to the WUF-W 

connection. Its performance is comparable to that of the WUF-W. This 

connection, rather than the cover-plated connection commonly used from 1994 

until publication of FEMA-267A, has been recommended for use in new 

buildings, because the welding of a single thickness of plate is considered to be 

more reliable than the welding of the combination of the beam flange and a 

cover-plate. 

A CJP groove welded web connection is required for use in this prequalified 

connection, since such a web connection was used in the tested connections. 

Tests using bolted webs have not been reported. 

The reader is referred to FEMA-355D, State Of the Art Report on Connection 

Performance, for more information on the testing and performance of this type of 

connection. 

3.5.4.1 Design Procedure 

Step 1: Select preliminary length of flange plate. 

Step 2: Choose the width of the flange plate, bp, based on beam flange width. 

Step 3: Calculate Mpr, Mc, and Myf according to Section 3.2.6. 

Step 4: Calculate tp based from the equation: 

M 

t p = 

yf (3-13) 

Fypbp �

�

�

Ł 

db + 

t pl b 

2 

+ t pl t �

ł

�

� 

where: 

bp = Width of flange plate at column face. Tapered plates should be checked for 

the critical section 

t

plt 

and tplb are the thicknesses of the top and bottom flange plates, respectively. 

Step 5: Calculate the length and thickness of the weld of the flange plate to the beam flange 

using the equation: 

M 

lwtw = 

f (3-14) 0.707Fw 

where: 

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lw = total length of weld including end weld (see Fig. 3-11). 

Fw = nominal design strength of weld from AISC-LRFD = 0.60FEXX 

tw (max) = t p - 1 

16 

inch 

If plate dimensions do not permit sufficient weld, return to Step 1 and select a longer 

plate length. 

Step 6: Determine the required panel zone thickness according to the methods of Section 

3.3.3.2. For purposes of this calculation, substitute db+(tplt+tplb) for db and the 

t plt + t plb quantity db - 2 

for db - tfb 

Step 7: Determine continuity plate requirements according to Section 3.3.3.1, For this 

purpose, use the plate width as the quantity bf. 

Step 8: Detail the connection as shown in Figure 3-11. 

3.5.5 Reduced Beam Section Connections 

This section provides procedures for design of fully restrained, Reduced Beam Section (RBS) 

connections. These connections utilize circular radius cuts in both top and bottom flanges of the 

beam to reduce the flange area over a length of the beam near the ends of the beam span. Welds 

of beam flanges to column are complete joint penetration groove welds, meeting the 

requirements of FEMA-353, Recommended Specifications and Quality Assurance Guidelines for 

Steel Moment Frame Construction for Seismic Applications. In this type of connection, no 

reinforcement, other than weld metal, is used to join the flanges of the beam to the column. Web 

joints for these connections may be either complete penetration groove welds, or bolted or 

welded shear tabs. Table 3-6 provides limitations and details of the prequalification. Figure 3-

12 provides typical details for this connection type. These connections are prequalified for use in 

Special Moment Frame or Ordinary Moment Frame systems within the limitations indicated in 

Table 3-6. When this type of connection is used, the elastic drift calculations should consider the 

effect of the flange reduction. In lieu of specific calculations, a drift increase of 9% may be 

applied for flange reductions ranging to 50% of the beam flange width, with linear interpolation 

for lesser values of beam flange reduction. 

Commentary: This type of connection has performed adequately in tests with 

both welded and bolted web connections. While a welded web connection is more 

costly than the more conventional bolted web connection, it is believed that the 

welded web improves the reliability of the connection somewhat. The welded web 

provides for more effective force transfer through the web connection, thereby 

reducing stress levels at the beam flanges and beam flange groove welds. 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

Notes 

1. See Section 3.5.5.1 for calculation of RBS dimensions. See FEMA-353, Recommended Specifications 

and Quality Assurance Guidelines for Steel Moment Frame Construction for Seismic Applications, for 

fabrication details including cutting methods and smoothness requirements. 

2. See Figure 3-8, and Note 1 to Figure 3-8, except that weld access hole may be as shown there, or as in 

AISC LRFD Vol. 1, Fig. C-J1.2, for rolled shapes or groove welded shapes. 

3. Web Connection: Erection bolts: number, type, and size selected for erection loads. 

a. Alternative 1: CJP welded web. Weld QC/QA Category BM/L. Shear tab length is equal to the 

distance between the weld access holes plus 1⁄4”. Shear tab thickness is as required for erection and 

the tab serves as backing for CJP weld (3/8” min. thickness). Shear tab may be cut square, or 

tapered as shown. Weld of shear tab to column flange is minimum 3/16” fillet on the side of the 

beam web, and a fillet sized for erection loads (5/16” minimum) on the side away from the beam 

web. No weld tabs are required at the ends of the CJP weld and no welding of the shear tab to the 

beam web is required.Weld: QC/QA Category BM/L. Erection bolts are sized for erection loads. 

b. Alternative 2: Bolted shear tab. Shear tab and bolts are sized for shear, calculated as in Section 3.2 

and using the methods of AISC. The shear tab should be welded to the column flange with a CJP 

groove weld or fillet of 3⁄4 tpl on both sides. Weld: QC/QA Category BL/T. Bolts shall be ASTM 

A325 or A490, and shall be fully-tightened. 

4. For continuity plates and web doubler plates see Figure 3-6. 

Figure 3-12 Reduced Beam Section (RBS) Connection 

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Table 3-6 Prequalification Data for RBS Connections 

General 

Applicable systems OMF, SMF 

Hinge location distance sh dc / 2 +a+b/ 2 

Critical Beam Parameters 

Depth range W36 and shallower (maximum weight 300 lbs/ft) 

Minimum span-to-depth ratio OMF: 5. 

SMF: 7 

bf /2tf Up to 52/� Fy, with bf determined as described in 

Section 3.3.1.1 

Flange thickness range 1-3/4” maximum 

Permissible material specifications A572 Grade 50, A992, A913 Grade 50/S75 

Flange reduction parameters Sec 3.5.5.1 

Critical Column Parameters 

Depth range OMF: Not Limited 

SMF: W12, W14 

Permissible material specifications A572 Grade 50; A913 Grade 50 and 65, A992 

Beam / Column Relations 

Panel Zone strength SMF: Section 3.3.3.2 

Column/beam bending strength ratio Section 2.9.1 

Connection Details 

Web connection Section 3.5.5.1 and Figure 3-12 

Continuity plate thickness Sec. 3.3.3.1 

Flange welds Fig. 3-12 

Welding parameters Sections 3.3.2.4, 3.3.2.5, 3.3.2.6 

Weld access holes See Fig. C-J1.2 AISC LRFD Vol. 1, or Section 

3.3.2.7 

As an alternative to a CJP groove weld, the beam web connection can also be 

made using a welded shear tab. The shear tab may be welded to the column using 

either fillet welds or groove welds. The shear tab, in turn, is then welded to the 

beam web with fillet welds. It is important to extend the tab as described in Figure 

3-12, so as not to cause stress concentration near the end of the weld access hole. 

The web connection can also be made with a shear tab that is welded to the 

column flange and bolted to the beam web. 

The effect of flange reduction on the elastic drift of frames can be readily 

calculated using prismatic beams with reduced moments of inertia or multisegment beams that accurately represent the reduced section properties. Studies 

have been performed at the University of Texas (Grubbs, 1997) that have shown 

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FEMA-350 Criteria for New Steel 

ranges of drift increase from 4% to 7%, depending on the amount of flange 

reduction and other factors. The default factor to increase drift is expected to be 

slightly conservative for most cases. 

3.5.5.1 Design Procedure 

Step 1: Determine the length and location of the beam flange reduction, based on the 

following: 

a @ (0.5 to 0.75) bf (3-15) 

b @ (0.65 to 0.85) db (3-16) 

where a and b are as shown in Figure 3-12, and bf and db are the beam flange width 

and depth respectively. 

Step 2: Determine the depth of the flange reduction, c, according to the following: 

a) Assume c = 0.20bf. 

b) Calculate ZRBS. 

c) Calculate Mf according to the method of Section 3.2.6 and Figure 3-4 using 

Cpr = 1.15. 

d) If Mf < CprRyZbFy the design is acceptable. If Mf is greater than the limit, increase 

c. The value of c should not exceed 0.25 bf. 

Step 3: Calculate Mf and Mc based on the final RBS dimensions according to the methods of 

Section 3.2.7. 

Step 4: Calculate the shear at the column face according to the equation: 

M f Vf = 2 

L - d 

+ Vg (3-17) 

c 

Where: Vg= shear due to factored gravity load. 

Step 5: Design the shear connection of the beam to the column. If a CJP welded web is used, 

no further calculations are required. If a bolted shear tab is to be used, the tab and 

bolts should be designed for the shear calculated in Step 4. Bolts should be designed 

for bearing, using a resistance factor f of unity. 

Step 6: Design the panel zone according to the methods of Section 3.3.3.2. 

Step 7: Check continuity plate requirements according to the methods of Section 3.3.3.1. 

Step 8: Detail the connection as shown in Figure 3-12. 

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3.5.5.2 Fabrication Requirements 

The RBS cut is normally made by thermal cutting. The finished cut should have a maximum 

surface roughness of 500 micro-inches, avoiding nicks, gouges, and other discontinuities. All 

corners should be rounded to minimize notch effects and cut edges should be ground in the 

direction of the flange length to have a surface roughness value as described in FEMA-353, 

Recommended Specifications and Quality Assurance Guidelines for Steel Moment Frame 

Construction for Seismic Applications. 

Commentary: Grinding parallel to the flange avoids grind marks perpendicular 

to the direction of stress, which can act as stress risers. It is not required to 

remove all vertical striations caused by flame cutting. 

3.5.5.3 Composite Construction 

When composite metal deck and concrete are used, welded studs should not be placed in the 

area of the beam flange between the column face and 6 inches beyond the extreme end of the 

RBS. See Section 3.3.1.6. 

3.6 Prequalified Bolted Fully Restrained Connections 

This section provides recommended criteria for alternative types of prequalified bolted, fully 

restrained, steel moment-frame connections suitable for use in new construction within the limits 

indicated in the prequalification for each detail. Table 3-7 indicates the various types of 

prequalified fully restrained connections, and the structural systems for which they are 

prequalified. Additional prequalification data on these various connection types is provided in 

the sections that follow. 

Table 3-7 Prequalified Bolted Fully Restrained Connections 

Connection Type Criteria Section Frame Type 

Bolted Unstiffened End Plate (BUEP) 3.6.1 OMF, SMF 

Bolted Stiffened End Plate (BSEP) 3.6.2 OMF, SMF 

Bolted Flange Plate (BFP) 3.6.3 OMF, SMF 

Double Split Tee (DST)* 3.7.1 OMF 

*This type of connection may be partially or fully restrained depending on design. 

3.6.1 Bolted Unstiffened End Plate Connections 

The bolted unstiffened end plate (BUEP) connection is made by shop welding the beam to an 

end plate using (1) a CJP welded joint of the beam flanges to the plate and (2) fillet welds for the 

beam web to the plate. The end plate is then field-bolted to the column. The CJP groove weld of 

the beam flange is made without using a weld access hole, and is therefore not a prequalified 

weld in the area of the beam web, where backing cannot be installed. However, qualification of 

this joint detail to meet AWS requirements is not necessary. This type of connection can be used 

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in either Ordinary Moment Frame or Special Moment Frame systems within the member size 

limitations given in Table 3-8. Figure 3-13 presents a detail for the connection. 

Notes 

1. ASTM A36 end plate. For sizing see Section 3.6.1.1. 

2. CJP groove weld. This weld has special requirements. See FEMA-353, Recommended Specifications and 

Quality Assurance Guidelines for Steel Moment Frame Construction for Seismic Applications, for 

fabrication details. Weld: QC/QA Category AH/T. 

3. Fillet weld both sides, or CJP weld; see Section 3.6.1.3 for sizing requirements. See FEMA-353, 

Recommended Specifications and Quality Assurance Guidelines for Steel Moment Frame Construction 

for Seismic Applications, for fabrication details. Weld: QC/QA Category BM/L. 

4. Pretensioned ASTM A325 or A490 bolts. Diameter not to exceed 1-1/2 inch. See Section 3.6.1.1 for 

sizing requirements. 

5. Bolt location is part of the end plate design. See Section 3.6.1.1. 

6. For continuity plates and web doubler plates, see Figure 3-6. For calculation of panel zone strength, see 

Section 3.6.1.1. 

7. Shim as required. Finger shims shall not be placed with fingers pointing up. 

Figure 3-13 Bolted Unstiffened End Plate (BUEP) Connection 

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Table 3-8 Prequalification Data for BUEP Connections 

General 

Applicable systems OMF, SMF 

Hinge location distance, sh dc /2 + tpl + db /3 

Critical Beam Parameters 

Maximum depth W30 and smaller for OMF 

W24 and smaller for SMF 

Minimum span-to-depth ratio OMF: 5 

SMF: 7 

Flange thickness Up to 3⁄4” 

Permissible material specifications A572 Grade 50, A992, A913 Grade 50/S75 

Critical Column Parameters 

Depth range OMF: Not limited 

SMF: W8, W10, W12, W14 

Flange thickness Section 3.6.1.1, Step 7 

Permissible material specifications A572, Grade 50; A913 Grade 50, or 65, A992 

Beam /Column Relations 

Panel zone strength Sec. 3.3.3.2, Section 3.6.1.1, Step 9. 

Column/beam bending strength ratio Sec. 2.9.1 

Connection Details 

Bolts: 

Bolt diameter Section 3.6.1.1, Step 2 

Bolt grades A325 & A490. 

Installation requirements Pretensioned 

Washers Single F436 when required. 

Hole type Standard 

End Plate: 

End plate thickness Section 3.6.1.1, Steps 3 and 4 

End plate material A36 

Flange Welds: 

Weld type CJP groove weld similar to AWS TC-U4b, 3/8” 

fillet used as backing, root backgouged prior to start 

of groove weld. See Fig. 3-13. 

Filler metal Section 3.3.2.4 

Weld access holes Not permitted 

Web connection: Figure 3-13 

Continuity plate thickness Section 3.6.1.1, Steps 6 and 8 

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Commentary: The behavior of this type of connection can be controlled by a 

number of different modes including flexural yielding of the beam section, flexural 

yielding of the end plates, yielding of the column panel zone, tension failure of the 

end plate bolts, shear failure of the end plate bolts, and failure of the various 

welded joints. Some of these modes are brittle, and therefore are undesirable, 

while others have significant ductility. Flexural yielding of the beam and shear 

yielding of the column panel zone are behavioral modes capable of exhibiting 

acceptable levels of inelastic behavior. Other modes are not. In order to design a 

connection of this type, it is necessary to select which modes of behavior are to be 

permitted to control the connection’s inelastic deformation. Once desired modes 

of behavior for the connection are selected, the various elements of the connection 

are designed with sufficient strength so that other modes are unlikely to occur. 

FEMA-355D, State of the Art Report on Connection Performance, provides 

further discussion of the performance of these connections, and summaries of test 

data and references. 

3.6.1.1 Design Procedure 

The connection shall be designed so that yielding occurs either as a combination of beam 

flexure and panel zone yielding or as beam flexure alone. The end plate, bolts and welds must be 

designed so that yielding does not occur in these elements.The design should be performed using 

the steps below. The various parameters used in the equations are defined in Figure 3-14 and in 

AISC-LRFD. 

Step 1: Calculate Mf and Mc according to the methods of Section 3.2.7. 

Step 2: Select end plate bolt size by solving Equation 3-20 for Tub and selecting bolt type and 

Abolt as required: 

M f < 2Tub (do + di ) (3-18) 

where: 

Tub = 90Abolt for A325 bolts 

= 113Abolt for A490 bolts 

and do and di are as defined in Figure 3-14 

Step 3: Check the adequacy of the selected bolt size to preclude shear failure by ensuring that 

the area Ab of the bolts satisfies the formula: 

2M f + V 

L - d g 

c Ab ‡ (3-19) 3Fv 

Step 4: Determine the minimum end plate thickness tp required to preclude end plate flexural 

yielding from the equation: 

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t p = 

( ) ( ) 1 1 2 0.8

2 2 

f

p b 

yp b t 

f 

M 

b d F d p p 

p s g 

� � Ø � � � � - Œ � + � + + + � � � � � � Œ o Ł ł Ł

1 

2 

p 

f 

f 

b 

s

p 

ø � �

� œ + � � � œß ł�� 

(3-20) 

where: 

s = b g p (3-21) 

g = is the bolt gage as defined in Figure 3-14

Note that the end plate is required to be ASTM A36 steel.

Step 5: Determine the minimum end plate thickness required to preclude end plate shear 

yielding from the equation: 

M f t = p 1.1Fypbp (db - tbf ) (3-22) 

Step 6: Determine the minimum column flange thickness required to resist beam flange 

tension from the equation: 

tfc = 2 

f 

1 

b fb 

yc 

M 

C 

dt 

Fc 

- (3-23) 

where: 

C1 = 

g - k1 (3-24) 2 

k1 = Distance from centerline of column web to flange toe of fillet as defined in 

AISC Manual. 

If the column flange thickness is less than the calculated requirement, continuity 

plates are required. Continuity plates, if required, shall be sized as required in 

Section 3.3.3.1. 

Step 7: If continuity plates are required, the column flange thickness must be additionally 

checked for adequacy to meet the following: 

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t fc > 

yc c 

b fb 

f

F Y 

d t 

M 

0.8 

2 ( - ) (3-25) 

where: 

Yc = �

� c 

+ s �

�

�

� 1 

+ 

2 �

� + (C2 + C1 )

�

� 42 �� + (3-26) Ł 2 łŁ C2 C1 ł Ł cs ł 

g C1 = - k1 (3-27) 2 

C= 

bfc - g (3-28) 2 2 

s = ( 24 ) 2

12 

fc 

2 

CC b k

C + 1 C1 

(3-29) 

If tc is less than the calculated value, a column with a thicker flange must be selected. 

Step 8: Check column flange thickness for adequacy for beam flange compression according 

to the following: 

M f t fc >

( db - t fb )(6k + 2t pl + tbf ) Fyc 

(3-30) 

where k is the k-distance of the column from the AISC Manual. 

If tfc is less than given by Equation 3-30, than beam flange continuity plates are 

required in accordance with Section 3.3.3.1. 

Step 9: Check the panel zone shear capacity in accordance with Section 3.3.3.2. For 

purposes of this calculation, db may be taken as the distance from one edge of the end 

plate to the center of the beam flange at the opposite flange. 

Step 10: Detail the connection as shown in Figure 3-13. 

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bp 

pf 

pf d1 

d0 

g 

bp 

pf 

pf d1 

d0 

bp 

pf 

pf 

tw 

d1 

d0 

g 

db 

tbf 

tpl 

db 

tbf 

tpl 

db 

tbf 

tpl 

Figure 3-14 Geometry of Unstiffened End Plate Connection 

3.6.2 Bolted Stiffened End Plate Connection 

This bolted stiffened end plate (BSEP) connection is made by shop-welding the beam to the 

end plate using (1) a CJP welded joint for the beam flanges to the end plate and (2) fillet welds 

for the beam web to end plate. The endplate is then field-bolted to the column. The CJP groove 

weld of the beam flange is made without using a weld access hole, and is therefore not a 

prequalified weld in the area of the beam flange, where backing cannot be installed. However, 

qualification of this joint detail to meet AWS requirements is not necessary. The outstanding 

flanges of the end plate at the top and bottom of the beam are stiffened by a vertical fin plate that 

extends outward from the beam flanges. These stiffener plates are CJP double-bevel groove 

welded to the beam flanges and end plates. This type of connection can be used in either 

Ordinary Moment Frame or Special Moment Frame systems within the limitations given in Table 

3-9. A detail of this connection type is shown in Fig. 3-15. 

Commentary: The behavior of this type of connection can be controlled by a 

number of different behavioral modes including flexural yielding of the beam 

section, flexural yielding of the end plates, yielding of the column panel zone, 

tension failure of the end plate bolts, shear failure of the end-plate bolts, and 

failure of the various welded joints. Some of these modes are brittle, and 

therefore are undesirable while others have significant ductility. Flexural 

yielding of the beam and shear yielding of the column panel zone are behavioral 

modes capable of exhibiting acceptable levels of inelastic behavior. Other modes 

are not. The design procedure contained in this section is based on inelastic 

action occurring in preferred modes. The various elements of the connection are 

then designed with sufficient strength so that other modes are unlikely to occur. 

FEMA-355D, State Of Art Report on Connection Performance, provides further 

discussion of the performance of these connections and summaries of test data 

and references. 

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Notes 

1. ASTM A36 end plate. For sizing, see Section 3.6.2.1. 

2. CJP groove weld. This weld has special requirements. See FEMA-353, Recommended Specifications and 

Quality Assurance Guidelines for Steel Moment Frame Construction for Seismic Applications, for 

fabrication details. Weld: QC/QA Category AH/T. 

3. Fillet weld both sides, or CJP weld; see Section 3.6.2.4 for sizing requirements. See FEMA-353, 

Recommended Specifications and Quality Assurance Guidelines for Steel Moment Frame Construction 

for Seismic Applications, for fabrication details. Weld: QC/QA Category BM/L. 

4. Pretensioned ASTM A325 or A490 bolts. See Section 3.6.2.1 for sizing requirements. 

5. Bolt location is part of the end plate design. See Section 3.6.2.1. 

6. For continuity plates and web doubler plates, see Figure 3-6. For calculation of panel zone strength, see 

Section 3.6.2.1. 

7. Stiffener is shaped as shown. Stiffener thickness shall be the same as that of the beam web. 

8. Stiffener welds are CJP double-bevel groove welds to both beam flange and end plate. Weld: QC/QA 

Category AH/T for weld to endplate. BM/L for weld to beam.. 

9. Shim as required. Finger shims shall not be placed with fingers pointing up. 

Figure 3-15 Stiffened End Plate Connection 

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Table 3-9 Prequalification Data for Bolted Stiffened End Plate Connections 

General 

Applicable systems OMF, SMF 

Hinge location distance sh dc /2 + tpl + Lst 

Critical Beam Parameters 

Maximum depth W36 

Minimum span-to-depth ratio OMF: 5 

SMF: 7 

Flange thickness 1” 

Permissible material specifications A572 Grade 50, A992, A913 Gr50/S75 

Critical Column Parameters 

Depth range OMF: Not Limited 

SMF: W12, W14 

Flange thickness Section 3.6.2.1, Step 6 

Permissible material specifications A572, Grade 50; A913 Grade 50 and 65, A992 

Beam /Column Relations 

Panel zone strength Sec. 3.6.2.1, Step 7 

Column/beam bending strength ratio Sec. 2.9.1 

Connection Details 

Bolts: 

Bolt diameter Section 3.6.2.1, Step 1 

Bolt grades A325 and A490. 

Installation requirements Pretensioned 

Washers Single F436 when required 

Hole type Standard 

End Plate: 

End plate thickness and rib size Section 3.6.2.1, Step 2 

End plate and rib material specification A36 

Flange welds: 

Weld type CJP groove weld similar to AWS TC-U4b, 3/8” 

fillet used as backing, root backgouged prior to start 

of groove weld. See Fig. 3-15. 

Weld metal Section 3.3.2.4 

Weld access holes Not permitted 

Web connection: Figure 3-15 

Continuity plate thickness Section 3.6.2.1, Steps 4 and 5 

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3.6.2.1 Design Procedure 

The connection shall be designed so that yielding occurs either as a combination of beam 

flexure and panel zone yielding or as beam flexure alone. The design should be performed using 

the steps below. The various parameters used in the equations are defined in Figure 3-16 and in 

AISC-LRFD. 

Step 1: Calculate Mf and Mc according to the methods of Section 3.2.7. 

Step 2: Select end plate bolt size by solving Equation 3-32 for Tub and selecting bolt type and 

Abolt as required: 

M f < 3.4Tub ( do + di ) (3-31) 

where: 

Tub = 90Abolt for A325 bolts 

= 113Abolt for A490 bolts 

and do and di are as defined in Fig. 3-16 

Confirm that Tub satisfies the Equation: 

0.00002305 p f 

0.591 (Ffu )2.583 

Tub ‡ 

t 0.895 d 1.909 t 0.327 b 0.965 + Tb (3-32) 

p bt s p 

Where Tb is the minimum bolt pretension per Table J3.1 of AISC-LRFD. 

Adjust bolt size as required. 

Step 3: Check the adequacy of the selected bolt size to preclude shear failure by ensuring that 

the area Ab of the bolts, satisfies the formula: 

2M f + V 

L - d g 

c Ab ‡ (3-33) 6Fv 

Step 4: Determine the minimum end plate thickness tp required to preclude end plate flexural 

yielding as the larger of the values given by equations 3-34 or 3-35: 

0.00609 p f 

0.9 

g 0.6 Ffu 

0.9 

t p ‡ 

d 0.9 t 0.1 b 0.7 (3-34) 

bt s p 

0.00413 p f 

0.25 

g 0.15 Ffu t p ‡ 

d 0.7 t 0.15 b 0.3 (3-35) 

bt s p 

where: 

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M 

Ffu = f (3-36) db -tbf 

and dbt is the diameter of the bolt 

Note that the end plate is required to be ASTM A36 steel and the stiffener plate must 

be at least as thick the beam web. 

Step 5: Determine the minimum column flange thickness required to resist beam flange 

tension from the equation: 

tcf > 

0.9 (3.5 ) 

( ) 3 

F p c

F C

yc b 

m fu 

+ 

a

(3-37) 

where: 

� Af a m = Ca �

�

Ł Aw ł

�

�

� 1/ 3 C3 

(dbt )

1/ 4 (3-38) 

C3 = 

g - dbt - k1 (3-39) 2 4 

and Ca = 1.45 for A325 bolts and 1.48 for A490 bolts when A36 end plates are used 

If the column flange is thinner than required, continuity plates are required and 

should be provided in accordance with Section 3.3.3.1. 

Step 6: Check column flange thickness for adequacy for beam flange compression according 

to the following: 

M f twc = (3-40) (db - t fb ) (6k + 2t p + t fb ) Fyc 

where k is the k-distance of the column from the AISC Manual. 

If the above relationship is not satisfied, continuity plates are required and should be 

provided in accordance with Section 3.3.3.1. 

Step 7: If continuity plates are required, the column flanges must be at least as thick as the 

required end plate thickness, calculated in Step 4. 

Step 8: Check the shear in the panel zone in accordance with Section 3.3.3.2. For purposes 

of this calculation, db may be taken as the distance from one end of the end plate to 

the center of the opposite flange. 

Step 9: Detail the connection as shown is Figure 3-15. 

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db 

tbf 

t b pl p 

pf 

pf 

tw 

d

d

1 

0 

g 

ts 

pb 

c pb 

30 o 

1ˆ 

1ˆ 

Figure 3-16 Geometry of Stiffened End Plate Connection 

3.6.3 Bolted Flange Plate Connections 

This section provides procedures for design of bolted flange plate (BFP) connections utilizing 

plates welded to the column flanges and bolted to the beam flanges. The flange plates are welded 

to the column flange using CJP welds following the recommendations given in sections 3.3.2.1 

through 3.3.2.5. The flange plates are bolted to beam flanges following the recommendations of 

Sections 3.3.4.1 and this Section. The beam web is connected to the column flange with a bolted 

shear tab. A detail for this connection type is shown in Figure 3-17. Table 3-10 presents the 

limitations for this connection prequalification. Figure 3-18 shows dimensions and nomenclature 

to be used with the design procedure of Section 3.6.3.1. 

Commentary: The behavior of this type of connection can be controlled by a 

number of different modes including: flexural yielding of the beam section, 

flexural yielding of the cover plates, yielding of the column panel zone, netsection tensile failure of the beam flange or cover plates, shear failure of the 

bolted connections, or failure of the welded joints. Some of these modes are 

brittle, while others have significant ductility. Connections of this type must be 

controlled by a preferred ductile behavior where the various elements of the 

connection are designed with sufficient strength that the other modes are unlikely 

to occur. Tests of connection assemblies incorporating this detail, as described in 

FEMA-355D, indicate that the best inelastic behavior is achieved with balanced 

yielding in all of the three preferred mechanisms: beam flexure, cover plate 

extension and compression, and panel zone yielding. When this balanced 

behavior occurs, the required rotations may be met without any of the 

mechanisms fully developing their maximum strain-hardened strength. For 

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Criteria for New Steel FEMA-350 

example, CprRyFyZ of the beam may not be reached at the beam yield section. For 

this reason, and unlike the case with some other prequalified connections, the 

design equations are developed at the onset of yielding, rather than at full yield. 

Notes 

1. Size the flange plate and bolts in accordance with Section 3.6.3.1. Bolts are fully pretensioned ASTM 

A325 or A490, designed for bearing. Bolt holes in flange plate are oversize holes. Use standard holes in 

beam flange. Washers as required by RCSC, Section 7. 

2. CJP groove weld, single or double bevel. Weld in shop or field. When using single-bevel groove weld, 

remove backing after welding, backgouge, and reinforce with 5/16” minimum fillet weld. When using 

double bevel weld, backgouge first weld before welding other side. Weld: QC/QA Category AH/T. 

3. Shims are permitted between flange plates and flanges. 

4. Size shear tab and bolts by design procedure in Section 3.6.3.2. Bolt holes in shear tab are short-slottedhorizontal; holes in web are standard. Weld QC/QA Category BM/L. 

5. For continuity plates and web doubler plates see Figure 3-6. For calculation of continuity plate requirements, 

use flange plate properties as flange properties. 

Figure 3-17 Bolted Flange Plate (BFP) Connection 

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Table 3-10 Prequalification Data for Bolted Flange Plate Connections 

General 

Applicable systems OMF, SMF 

Hinge location distance sh dc /2 + Lp 

Critical Beam Parameters 

Maximum depth OMF: up to W36 

SMF: up to W30 

Minimum span-to-depth ratio OMF: 5 

SMF: 8 

Flange thickness Up to 1-1/4” (OMF) 

Up to 3⁄4” (SMF) 

Permissible material specifications A572 Grade 50, A992, A913 Gr50/S75 

Critical Column Parameters 

Depth range OMF: Not Limited 

SMF: W12, W14 

Permissible material specifications A572 Grade 50, A913 Grade 50 or 65, A992 

Critical Beam Column Relations 

Panel zone strength Section 3.6.3.1, Step 3. 

Column/beam bending strength ratio Section 2.9.1 

Critical Connection Details 

Connection Plates: 

Permissible material specifications A36, A572 Grade 42 or 50 

Design method Section 3.6.3.1, Step 4 and Step 5 

Weld to flange Fig. 3-17. Welding QC/QA Category AH. 

Flange welding parameters Section 3.3.2.4, 3.3.2.5, 3.3.2.6 

Bolt Characteristics: 

Bolt diameter Section 3.6.3.1, Steps 6 and 7; 1-1/8” maximum 

Bolt grade A325-X or A490-X 

Bolt spacing 3x bolt diameter min. 

Installation requirements Pretensioned 

Washers F436 as required 

Web Connection Parameters: 

Web Connection Section 3.6.3.1, Step 12; Shear tab welded to column 

flange and bolted to beam. Bolt holes short-slotted 

horizontal. See Fig. 3-17. 

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db 

Tpl−t Tpl−b 

s 

bf 

c 

s2 

N= Number of flange bolts in row 

dc 

db 

Tpl−t Tpl−b 

s1 s3=(N−1)s2 s4 

b

g f 

c 

twc 

bp 

s2 

Figure 3-18 Geometry of the Bolted Flange Plate Connection 

3.6.3.1 Design Procedure 

The design of the connection should be performed using the steps below. The various 

parameters used in the equations are defined in Figure 3-18, and in AISC/LRFD. 

Step 1: Calculate Mf and Mc according to the procedures in Section 3.2.7. 

Step 2: Calculate the moment at the face of the column at onset of beam flange yielding, Myf, 

according to Section 3.2.8. 

Step 3: Calculate panel zone thickness requirements according to Section 3.3.3.2. It is 

recommended not to overstrengthen the panel zone for these connections. If the 

thickness of the panel zone is more than 1.5 times that required, it is recommended to 

use a different combination of beam and column sizes. Use the distance between the 

outer faces of the flange plates as db, and the center-to-center distance between plates 

in place of the quantity "db - tf " in the application of the procedure of Section 3.3.3.2. 

Step 4: Establish the width of the flange plate, bp, based on the geometry of the beam and 

column. 

Step 5: Calculate the minimum required thickness of the flange plates, tpl from the equation: 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

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t pl 

d b -

= 

4.4

2 

y p 

f y 

b F b 

M 

d -

(3-41) 2 

Commentary: It is desirable not to oversize the flange plates, as best performance 

is achieved with a combination of yielding of the beam flange, the panel zone, and 

the flange plate. 

Step 6: Select the number, size and grade of bolts in the beam-flange-to-flange-plate 

connection and evaluate the adequacy of the plate and beam, to preclude net section 

failures and bolt hole elongation failures, in accordance with Steps 7, 8, 9, 10 and 11 

respectively,where in each case, Equation 3-42 must be satisfied: 

1.2M y f < M fail (3-42) 

where: 

Myf = Moment at the face of the column at initiation of beam flange yielding, 

calculated in Step 2 above, and 

Mfail= Moment at the face of the column at initiation of failure in the specific 

behavior mode being addressed in Steps 7 through 11. 

Step 7: Determine Mfail the moment at the face of the column for shear failure of the bolts in 

accordance with Equation 3-43 and check for adequacy to meet the criteria of 

Equation 3-42, Step 6: 

M fail 2NAb ( Fv bolts = bolt ) db LTF1 (3-43)

where: 

Ab = Area of bolt 

Fvbolt = Nominal shear strength of bolt in bearing-type connections, from AISC 

LRFD. 

LTF1 = Length ratio to transfer moment from center of bolt group to face of 

column given by Equation 3-44: 

L d L = - c 

TF1 L d S S (3-44) - c - (2 1 + 3 ) 

N = Number of bolts in connection of beam flange to flange plate 

Step 8: Determine Mfail the moment at the face of the column for net section fracture of the 

flange plate in accordance with Equation 3-45 and check for adequacy to meet the 

criteria of Equation 3-42, in Step 6: 

M fail 0.85 Fu pl (bp 2(dbt 0 .062)) t pl (db t pl = - + + ) LTF 2 (3-45) FP - hole 

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where: 

dbthole = diameter of flange plate bolt hole, inches. 

LTF2 = ratio to transfer moment at bolt hole closest to column to column face given by 

Equation 3-46: 

L d - c LTF2 = (3-46) L - d - 2S c 1 

Step 9: Determine Mfail the moment at the face of the column for net section fracture of the 

beam flange in accordance with Equation 3-47 and check for adequacy to meet the 

criteria of Equation 3-42, Step 6: 

M fail = Fu-b ( Zb - 2 (dbt hole + 0.062) t fb ( db - t fb )) LTF3 (3-47) 

where: 

dbt = diameter of bolt, inches 

LTF3 = ratio to transfer moment from the bolt hole furthest from the column face to the 

column face, given by Equation 3-48: 

L d - c LTF3 = (3-48) L - dc - 2(S1 + S3 ) 

Step 10: Determine Mfail the moment at the face of the column for elongation of bolt holes in 

accordance with Equation 3-49 and check for adequacy to meet the criteria of 

Equation 3-42, Step 6: 

' 

M -

fail = Tn �

� 

db + 

tPL-t + tPLb �

� LTF1 (3-49) 

Ł 2 ł 

where: 

Tn is the lesser of the values given by Equations 3-50 or 3-51: 

Tn = 2.4Fu -b ( S3 + S1 - c tfb ) (3-50)

Tn = 2.4Fu - pl ( S3 + S t pl (3-51) 4 )

Step 11: Check block shear according to the requirements of AISC LRFD to ensure that the 

moment at the column face due to any of these modes meets the requirements of the 

relationship in Step 6. The block shear failure modes are shown in Figure 3-19. For 

the purpose of this calculation, the resistance factor f shall be taken as unity. 

Step 12: Design a single-plate, bolted shear-tab connection sufficient to resist the shear given 

by Equation 3-52: 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

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2M 

Vweb = 

f + V (3-52) L - d g 

c 

where Vg is the shear at the column face due to factored gravity loads, kips. 

Step 13: Calculate continuity plate requirements in accordance with the methods of Section 

3.3.3.1 using the width and thickness of the flange plates for the quantities bf, and tf, 

respectively, in that section. 

Step 14: Confirm the adequacy of the column size to meet the criteria of Section 2.8.1, 

considering the hinge location given in Table 3-10. 

Step 15: Detail the connection as shown in Fig. 3-17. Bolts should be designed for bearing using 

a resistance factor f of unity. 

Block Shear of Tension Edge Block Shear of 

Flange Plate Flange Plate 

Block Shear of Beam 

Flange 

Bolt Pull Through for Bolt Pull Through for 

Flange Plate Beam Flange 

Figure 3-19 Block Shear and Pull-Through Failures 

3.7 Prequalified Partially Restrained Connections 

This section provides recommended criteria for one type of prequalified full strength / partial 

stiffness (Partially Restrained (PR)) steel moment-frame connection, suitable for use in new 

construction. Table 3-11 indicates the connection type, and the structural systems for which it is 

prequalified. A procedure is also provided for design of this connection type. 

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Table 3-11 Prequalified Bolted Partially Restrained Connections 

Connection 

Type 

Criteria 

Section 

Frame Type 

DST 3.7.1 OMF, SMF 

Commentary: Several different types of partially restrained connections have 

been studied under the program of investigations performed in support of the 

development of these Recommended Criteria as well as under other programs. In 

preparing these Recommended Criteria, it was judged that sufficient supporting 

data had been developed for only one connection type to permit prequalification. 

This is not to suggest that the other types of connections studied are not suitable 

for use, but rather, that in the judgment of the project team, there were not 

sufficient data available for these other connection types to meet all of the 

requirements for prequalification in these Recommended Criteria, as described in 

Section 3.4. For additional information on other types of PR connections refer to 

FEMA-355D, State of the Art Report on Connection Performance. 

For the purposes of this document, connections are classified as partial 

stiffness, or Partially Restrained (PR) if the deformation of the connection itself 

will increase the calculated drift of the frame by more than 10%. The connection 

is considered to be a full strength connection when it can develop the full 

expected plastic moment of the beam itself. 

Moment connections that develop only partial strength, as well as partial 

stiffness, are sometimes used. Such connections are not prequalified in this 

document. This is not intended to preclude their use, if the system is properly 

justified by both analysis and testing. A significant amount of testing does exist 

on such connection types, and is described in FEMA-355D. 

3.7.1 Double Split Tee Connections 

This section provides procedures for design of full-strength, partially restrained, double split 

tee (DST) connections employing bolted split tee connectors between the beam and column 

flanges. This type of connection is prequalified for use within the limitations indicated in Table 

3-12. Figure 3-20 provides a typical detail for this connection type. 

Commentary: The behavior of this type of connection can be controlled by a 

number of different modes including flexural yielding of the beam section, flexural 

yielding of the tee stems or flanges, shear yielding of the column panel zone, netsection tensile failure of the beam flange or tee stem, and shear or tension failure 

of the bolts, depending on the relative proportions of these various components. 

Some of these modes are brittle, while others have significant ductility. The 

design procedure contained in this Section is based on inelastic action occurring 

in preferred modes. The various elements of the connection are then designed 

with sufficient strength so that other modes are unlikely to occur. FEMA-355D, 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

State of the Art Report on Connection Performance, provides further discussion of 

the performance of these connections and summaries of test data and references. 

Notes 

1. Split Tee: length, width, and thickness determined by design according to Section 3.7.1.2. 

2. Fully pretensioned ASTM A325 or A490 bolts in standard holes sized for bearing. For sizing, see 

Section 3.7.1.2, Step 7. 

3. Fully pretensioned ASTM A325 or A490 bolts in standard holes sized for bearing. For sizing, see 

Section 3.7.1.2, Step 4. 

4. Shear tab welded to column flange with either CJP weld or two-sided fillet weld. For calculation of 

design strength of shear tab, welds, and bolts, see Section 3.7.1.2, Step 14. Weld: QC/QA Category 

BM/L. 

5. For continuity plates and web doubler plates see Figure 3-6. 

Figure 3-20 Double Split Tee (DST) Connection 

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Moment-Frame Buildings Chapter 3: Connection Qualification

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Table 3-12 Prequalification Data for Full Strength DST Connections (FSDST) 

General 

Applicable systems OMF, SMF 

Connection classification Full-Strength – Partial-Stiffness (PR) 

Hinge location distance sh End of T-stubs 

Critical Beam Parameters 

Maximum depth OMF: W36 

SMF: W24 

Minimum span-to-depth ratio OMF: 5 

SMF: 8 

Permissible material specifications A572 Grade 50, A992, A913, Grade 50/S75 

Critical Column Parameters 

Depth range OMF: Not Limited. 

SMF: W12, W14 

Flange width governed by required length of T-stub 

flange 

Permissible material specifications A572 Grade 50, A913 Grade 50 or 65, A992 

Flange thickness Section 3.7.1.2, Steps 11 and 12. 

Critical Beam Column Relations 

Panel zone strength Section. 3.7.1.2, Step 3 

Column/beam bending strength ratio Sec. 2.9.1 

Critical Connection Details 

T-stub Parameters: 

Hole type Standard 

Permissible material specifications A572 Grade 50, A992 

Design method 3.7.1.2 

Web connection parameters: 

Shear tab: 

Permissible material specifications A36, A572 Grade 50 

Plate thickness 5/16” to 1⁄2” 

Hole type SSLT 

Weld type CJP groove or double fillet. See Fig. 3-20. 

Weld metal Section 3.3.2.4 

Double web angle: 

Permissible material specifications A36, A572 Grade 50 

Angle thickness 5/16” to 1⁄2” 

Hole type STD, SSLT 

Bolt Characteristics: 

Bolt diameter 7/8” or 1” 

Bolt grade A325-X or A490-X 

Bolt spacing 3x bolt diameter min. 

Installation requirements Pretensioned 

Washers F436 as required 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

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3.7.1.1 Connection Stiffness 

The analysis of frames incorporating partially restrained connections must include explicit 

consideration of the stiffness of the connection. Stiffness of double split tee connections may be 

calculated from Equation 3-53: 

db M fail k = (3-53) s 0.375

where: 

ks = the rotational stiffness of the connection, kip-inches/radian 

Mfail = the lesser of the moments that control the resistance of the connection, taken from 

steps 4 thru 8. 

db = the beam depth, inches 

or more accurate representations of stiffness may be used, when substantiated by rational analysis 

and test data. 

Commentary: The stiffness of partially restrained connections, by definition, has 

significant effect on the total drift experienced by the building frame in response 

to lateral loading. In order to account properly for this effect it is necessary to 

include consideration of the connection stiffness in the analytical model used to 

determine building drift and the distribution of the forces on the members of the 

steel moment frame. This can be done either by explicitly including an element in 

the structural model with this calculated stiffness, or alternatively by modifying 

the stiffness of the beams in the model to include the effect of connection stiffness. 

Guidelines for both approaches are contained in Section 4.5.2.2. 

3.7.1.2 Design Procedure 

The connection shall be designed so that inelastic behavior is controlled either by flexural 

yielding of the beam in combination with shear yielding of the column panel zone, or by flexural 

yielding of the beam alone. The various elements of the connection are proportioned such that the 

moment at the face of the column, as limited by the controlling yielding behavior considering 

potential material over-strength and strain hardening, is less than the moment at the face of the 

column corresponding with failure of any of the other behavioral modes. 

Based on the above, the design should proceed following the steps below. The various 

parameters used in the equations in these steps are defined in Figure 3-21 and 3-22, and in 

AISC/LRFD. 

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Moment-Frame Buildings Chapter 3: 

3-65

Q Q gt

gt

B ’ B

b

b’

a’

a

r

0.5r

Ts,eff

Q Q gt

gt

B ’ B

b

b’

a’

a

r

0.5r

Ts,eff

Figure 3-21 Geometry for Prying Forces and Bending of T-Section Flanges

N= Number of flange bolts in row

dc

db

s1 s3=(N−1)s2 s4

b

gf

twc

w

s2

w1

o

o

tf−t

Tstem

Tstem

N= Number of flange bolts in row

dc

db

s1 s3=(N−1)s2 s4

b

gf

twc

w

s2

w1

o

o

tf−t

N= Number of flange bolts in row

dc

db

s1 s3=(N−1)s2 s4

b

gf

twc

w

s2

w1

o

o

tf−t

Tstem

Tstem

Figure 3-22 Geometry for Other T-Stub Failure Modes

Connection Qualification
Recommended Seismic Design 

Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

Step 1: Calculate Mf and Mc according to the procedures in Section 3.2.7. 

Step 2: Calculate the moment at the face of the column at beam first yield, Myf, in 

accordance with the procedures of Section 3.2.8. 

Step 3: Check the adequacy of the column for panel zone shear in accordance with the 

procedures of Section 3.3.3.2. For purposes of this calculation, db may be assumed 

to be equal to the distance from the outer end of the tee flange at one beam flange to 

the bottom of the opposite T stem, and the quantity "db-tfb" can be taken as the 

quantity db, described above, minus tstem/2. If the thickness of the panel zone is 

more than 1.5 times that required, it is recommended to use a different combination 

of beam and column sizes. 

Step 4: Select the number, size, and grade of bolts in the beam flange and WT or ST flange. 

Select the size of the WT or ST, and evaluate the adequacy of the bolts, plate and 

beam to preclude brittle failure modes in accordance with Steps 5 through 10, where 

in each case, Equation 3-54 must be satisfied: 

1.2M yf < M fail (3-54) 

where: 

Myf = Moment at the face of the column at initiation of beam flange yielding, 

calculated in Step 2 above, and 

Mfail = Moment at the face of the column at initiation of failure in the specific 

behavior mode being addressed in Steps 5 through 12. 

Step 5: Determine Mfail the moment at the face of the column at shear failure of the beam 

flange bolts in accordance with Equation 3-55 and check for adequacy to meet the 

criteria of Equation 3-54, in Step 4: 

M failbolts = 2NAb ( Fvbolt ) d LTF1 (3-55) b 

where: 

Ab = Area of bolt 

Fvbolt = Nominal shear strength of bolt in bearing-type connections, from AISC 

LRFD. 

LTF1 = Length ratio to transfer moment from center of bolt group to face of 

column given by the Equation: 

L - dc L TF1 = (3-56) L - dc - (2S1 + S3 ) 

N = Number of bolts in the connection of the beam flange to the flange plate 

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Moment-Frame Buildings Chapter 3: Connection Qualification

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Step 6: Determine the moment Mfail at the face of the column at net section fracture of the T 

stem in accordance with Equation 3-57 and check for adequacy to meet the criteria 

of Equation 3-54 in Step 4: 

M fail = FuT ( w - 2 (dbt + .125)) tstem [db + tstem ] LTF2 (3-57) 

where: 

w is taken as the lesser of the flange length of the T, the width of the T at the first 

line of bolts, as defined in Figure 22, or the quantity given by the equation: 

wgS3 tanqeff £+ (3-58)

o o 15 £ q eff = 60tstem £ 30 (3-59) 

LTF2 is a ratio to transfer moment from the center line of the bolts closest to the 

column flange to the face of the column, and is given by the equation: 

L d - c LTF2 = (3-60) L - d c - 2S1 

Step 7: Determine the moment Mfail at the face of the column at initiation of plastic bending 

of the tee flanges in accordance with Equation 3-61 and check for adequacy to meet 

the criteria of equation 3-54 in Step 4: 

� 2a¢ - d 

4 

bt 

ł

�

� w FyT t 2 

f (db - tstem ) 

¢ + 

M fail = Ł 

4a¢b¢ - dbt (b a¢) (3-61) 

where: 

d 

a¢ =+ bt a (3-62) 2 

d ¢ = - bt bb (3-63) 2 

Step 8: Determine the moment Mfail at the face of the column at the initiation of tensile 

failure of the bolts at the tee flange, considering prying action, in accordance with 

Equation 3-64 and check for adequacy to meet the criteria of equation 3-54 in 

Step 4: 

Ø yt M fail = Ntb (db + tstem ) ŒTub + 

wFt 2 

f ø

œ a¢ 

(3-64) ¢+ o 16a¢ ß a b¢ 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

where Tub is the nominal tensile strength of bolts from the T flanges to the column 

flange which should be taken as the quantity 90Abolt for A325 bolts and 113Abolt for 

A490 bolts. 

Step 9: Determine the moment Mfail at the face of the column at net section fracture of the 

beam flange, in accordance with Equation 3-65 and check for adequacy to meet the 

criteria of equation 3-54 in Step 4: 

M fail = {Fubm ( Zb - 2 ( dbt hole + 0.062) t fb ( db - t fb )) } LTF3 (3-65) 

where: 

LTF3 is a length ratio to transfer moment from the bolt hole farthest from the column 

face, to the column face, given by Equation 3-66: 

L - dc LTF3 = (3-66) L - dc - 2(S1 + S3 ) 

Step 10: Determine the moment Mfail at the face of the column at initiation of block shear 

failure and pull-through patterns of the stem of the tee (See Figure 3-19), according 

to the methods in AISC-LRFD. 

Step 11: Calculate the adequacy of column flange thickness for beam flange tension, in 

accordance with the equation: 

tcf ‡ 1.5t f -t (3-67) 

If the column flange thickness is less than that calculated in accordance with 

Equation 3-67, continuity plates are required. Continuity plates should be designed 

as described in Section 3.3.3.1. 

Step 12: Calculate the adequacy of column web thickness for the beam flange compression 

forces, in accordance with the equation: 

M f twc ‡ (db - tstem )(6k + c) Fyc 

(3-68) 

where k is the dimension of the column-flange-to-web fillet, as indicated in AISC 

Manual. 

If the column web thickness does not meet the criteria of Equation 3-68, then 

provide continuity plates in accordance with the criteria of Section 3.3.3.1. 

Step 13: If continuity plates are required, the column flange thickness must be equal to or 

larger than the flange thickness, tft, of the T. If the column flange thickness is less 

than this amount, a column with a thicker flange must be selected. 

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Step 14: Design the shear connection between the beam web and column as a standard shear 

tab welded to the column and bolted to the beam. Bolts shall be sized for bearing 

using a resistance factor f of unity. Design load for the shear tab shall be taken as 

given by the equation: 

M f Vst = 2 

L - d 

+ Vg (3-69) 

c 

where: 

Vst = Design shear force for the shear tab 

Vg = Factored gravity load 

Step 15: Detail the connection as shown in Figure 3-20. 

3.8 Proprietary Connections 

This section presents information on several types of fully restrained connections that have 

been developed on a proprietary basis. These connections are not categorized in these 

Recommended Criteria as prequalified, as the SAC Joint Venture has not examined the available 

supporting data in sufficient detail to confirm that they meet appropriate prequalification criteria. 

However, these proprietary connections have been evaluated by some enforcement agencies and 

found to be acceptable for specific projects and in some cases for general application within the 

jurisdiction’s authority. Use of these technologies without the express permission of the licensor 

may be a violation of intellectual property rights, under the laws of the United States. 

Discussion of several types of proprietary connections are included herein. Other proprietary 

connections may also exist. Inclusion or exclusion of proprietary connections in these 

Recommended Criteria should not be interpreted as either an approval or disapproval of these 

systems. The descriptions of these connections contained herein have in each case been prepared 

by the developer or licensor of the technology. This information has been printed with their 

permission. Neither the Federal Emergency Management Agency nor the SAC Joint Venture 

endorses any of the information provided or any of the claims made with regard to the attributes 

of these technologies or their suitability for application to specific projects. Designers wishing to 

consider specific proprietary connections for use in their structures should consult both the 

licensor of the connection and the applicable enforcement agency to determine the applicability 

and acceptability of the individual connection for the specific design application. 

3.8.1 Side Plate 

The proprietary side plate (SP) connection system is a patented technology shown 

schematically in Figure 3-23 for new construction. Physical separation between the face of 

column flange and end of beam eliminates peaked triaxial stress concentrations. Physical 

separation is achieved by means of parallel full-depth side plates that eliminate reliance on 

through-thickness properties and act as discrete continuity elements to sandwich and connect the 

beam and the column. The increased stiffness of the side plates inherently stiffens the global 

frame structure and eliminates reliance on panel zone deformation by providing three panel zones 

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Chapter 3: Connection Qualification Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

[i.e., the two side plates plus the column’s own web]. Top and bottom beam flange cover plates 

are used, when dimensionally necessary, to bridge the difference between flange widths of the 

beam(s) and the column. 

This connection system uses all shop fillet-welded fabrication. All fillet welds are made in 

either the flat or horizontal position using column tree construction. Shop fabricated column 

trees and link beams are erected and joined in the field using one of four link beam splice options 

to complete the moment-resisting frame. Link beam splice options include a fully welded CJP 

butt joint, bolted matching end plates, fillet-welded flange plates, and bolted flange plates. 

Figure 3-23 Proprietary Side Plate Connection 

All connection fillet welds are loaded principally in shear along their length. Moment 

transfer from the beam to the side plates, and from the side plates to the column, is accomplished 

with plates and fillet welds using equivalent force couples. Beam shear transfer from the beam’s 

web to the side plates is achieved with vertical shear plates and fillet welds. The side plates are 

designed with adequate strength and stiffness to force all significant plastic behavior of the 

connection system into the beam, in the form of flange and web local buckling centered at a 

distance of approximately 1/3 the depth of the beam away from the edge of the side plates. 

All full-scale cyclic testing of this connection system was conducted at the Charles Lee 

Powell Structural Research Laboratories, University of California, San Diego, under the direction 

of Professor Chia-Ming Uang. Testing includes both prototype uniaxial and biaxial dual strong 

axis tests. Independent corroborative nonlinear analyses were conducted by the University of 

Utah and by Myers, Houghton & Partners, Structural Engineers. 

Independent prequalification of this connection system was determined by ICBO Evaluation 

Service, Inc., in accordance with ICBO ES Acceptance Criteria for Qualification of Steel 

Moment Frame Connection Systems (AC 129-R1-0797), and was corroborated by the City of Los 

Angeles Engineering Research Section, Department of Building and Safety, which collectively 

invoke the qualification procedures contained in: FEMA 267/267A/267B; AISC Seismic 

Provisions for Structural Steel Buildings, dated April 15, 1997; and County of Los Angeles 

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Moment-Frame Buildings Chapter 3: Connection Qualification

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Current Position on Design and Construction of Welded Moment Resisting Frame Systems CP-2, 

dated August 14, 1996. Refer to ICBO Evaluation Service, Inc., Evaluation Report No. 5366, 

issued January 1, 1999, and to City of Los Angeles Research Report: COLA RR 25393 for 

allowable values and/or conditions of use. Additional independent jurisdictional scrutiny of this 

connection system, by Karl H. Frank, Ph.D., Egor P. Popov, Ph.D., C. Mark Saunders, S.E., and 

Robert L. Schwein, P.E. is contained in the Los Angeles County Technical Advisory Panel 

(LACO-TAP) SMRF Bulletin No. 3 on Steel Moment-Resisting Frame Connection Systems, 

County of Los Angeles, Department of Public Works, dated March 4, 1997. Additional design 

information for this connection type may be obtained from the licensor. 

3.8.2 Slotted Web 

The proprietary Slotted Web (SW) connection (Seismic Structural Design Associates, Inc. 

US Patent No. 5,680,738 issued 28 October 1997) is shown schematically in Figure 3-24. It is 

similar to the popular field-welded–field-bolted beam-to-column moment-frame connection, 

shown in the current AISC LRFD and ASD steel design manuals, that has become known as the 

“pre-Northridge” connection. Based upon surveys of seismic connection damage, modes of 

fracture, reviews of historic tests, and recent ATC-24 protocol tests, it was concluded by SEAOC 

(1996 Blue Book Commentary) that the pre-Northridge connection is fundamentally flawed and 

should not be used in the new construction of seismic moment frames. Subsequent finite element 

analyses and strain gage data from ATC-24 tests of this pre-Northridge connection have shown 

large stress and strain gradients horizontally across and vertically through the beam flanges and 

welds at the face of the column. These stress gradients produce a prying moment in the beam 

flanges at the weld access holes and in the flange welds at the column face that lead to beam 

flange and weld fractures and column flange divot modes of connection fracture. Moreover, 

these same studies have also shown that a large component, typically 50%, of the vertical beam 

shear and all of the beam moment, is carried by the beam flanges/welds in the pre-Northridge 

connection. 

However, by (1) separating the beam flanges from the beam web in the region of the 

connection and (2) welding the beam web to the column flange, the force, stress and strain 

distributions in this field-welded-field-bolted connection are changed dramatically in the 

following ways: 

1. The vertical beam shear in the beam flanges/welds is reduced from typically 50% to typically 

3% so that essentially all vertical shear is transferred to the column through the beam web 

and shear plate. 

2. Since most W-sections have a flange-to-beam modulus ratio of 0.65 < Zflg /Z < 0.75, both the 

beam web and flange separation and the beam web-to-column-flange weldment force the 

beam web to resist its portion of the total beam moment. 

3. The beam web separation from the beam flange reduces the large stress and strain gradients 

across and through the beam flanges by permitting the flanges to flex out of plane. Typically, 

the elastic stress and strain concentration factors (SCFs) are reduced from 4.0 to 5.0 down to 

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1.2 to 1.4, which dramatically reduces the beam flange prying moment and the accumulated 

plastic strain and ductility demand under cyclic loading. These attributes enhance and extend 

the fatigue life of this moment frame connection. 

4. The lateral-torsional mode of beam buckling that is characteristic of non-slotted beams is 

circumvented. The separation of the beam flanges and beam web allow the flanges and web 

to buckle independently and concurrently, which eliminates the twisting mode of buckling 

and its associated torsional beam flange/weld stresses. Elimination of this buckling mode is 

particularly important when the exterior cladding of the building is supported by seismic 

moment frames that are located on the perimeter of the building. 

5. Residual weldment stresses are significantly reduced. The separation of the beam web and 

flanges in the region of the connection provides a long structural separation between the 

vertical web and horizontal flange weldments. 

Figure 3-24 Proprietary Slotted Web Connection 

The SW connection design rationale that sizes the beam/web separation length, shear plate 

and connection weldments, is based upon ATC-24 protocol test results and inelastic finite 

element analyses of the stress and strain distributions and buckling modes. Incorporated in this 

rationale are the UBC and AISC Load and Resistance Factor Design (LRFD) Specifications and 

the AISC Seismic Provisions for Structural Steel Buildings. 

SSDA has successfully completed ATC-24 protocol tests on beams ranging from W27x94 to 

W36x280 using columns ranging from W14x176 to W14x550. None of these assemblies 

experienced the lateral-torsional mode of buckling that is typical of non-slotted beam and column 

assemblies. 

Both analytical studies and ATC-24 protocol tests have demonstrated that the Seismic 

Structural Design Associates Slotted Web connection designs develop the full plastic moment 

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capacity of the beam and do not reduce the elastic stiffness of the beam. All of the above 

attributes of this proprietary connection enhance its strength and ductility, which makes it 

applicable for use in new construction in seismic moment frames. Specific qualification and 

design information for the Slotted Web connection may be obtained from the licensor. 

3.8.3 Bolted Bracket 

The Bolted Bracket (BB) connection type is shown schematically in Figure 3-25. Beam shear 

and flexural stresses are transferred to the column through a pair of heavy, bolted brackets, 

located at the top and bottom beam flanges. The concept of using bolted brackets to connect 

beams to columns rigidly is within the public domain. However, generic prequalification data 

have not been developed for this connection type. One licensor has developed patented steel 

castings of the bolted brackets, for which specific design qualification data has been prepared. 

Specific qualification and design information for this connection type may be obtained from the 

licensor. 

Figure 3-25 Bolted Bracket Connection 

3.8.4 Reduced Web 

The reduced web (RW) section utilizes capacity design principles to protect the beam column 

connection from high stresses by introducing large openings in the web. The openings are large 

enough to cause yielding of the web along the beam span, allowing the connection region to 

remain nominally elastic. The configuration of openings can be adjusted to control the yielding 

mechanism and yield strength. Two configurations are illustrated in Figure 3-26. 

An understanding of the utility of this system for resisting seismic actions is developing. At 

this writing, five W21x68 Grade 50 beams have been tested under reversed cyclic loading using 

the modified SAC loading protocol. Stable hysteretic loops were maintained to interstory drifts 

as high as 6%, and the predicted deformation mechanisms developed. Modified pre-Northridge 

details, consisting of a field-bolted web connection and full penetration flange welds were shown 

to be successful. These followed the detailing recommended for the WUF-B connection except 

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the web was not welded. Use of the beams may reduce construction cost if mechanical equipment 

is passed through the openings, thereby allowing story heights to be reduced. The technology is 

protected by US Patent 6,012,256; inquiries are welcome. 

Figure 3-26 Reduced Web Connection 

3.9 Project-Specific Connection Qualification 

This section provides criteria for design and project-specific qualification of connections for 

which there is no current prequalification or for prequalified connections that are to be utilized 

outside the parametric limitations for the applicable prequalification. Project-specific 

qualification includes a program of connection assembly prototype testing supplemented by a 

suitable analytical procedure that permits prediction of behaviors identified in the testing 

program. 

Commentary: While it is not the intent of these Recommended Criteria to require 

testing for most design situations, there will arise circumstances where none of 

the prequalified connections will be appropriate, or where a prequalified 

connection must be used outside the parametric limits for which it is prequalified. 

In these situations, these criteria recommend a program of prototype testing in 

addition to analytically based connection design, reflecting the view that the 

behavior of connection assemblies under severe cyclic loading cannot be reliably 

predicted by analytical means alone. The program of laboratory testing is used to 

demonstrate that the behavioral modes of the connection are predictable and that 

the connection assembly is capable of adequate performance. The testing is 

accompanied by an analytical procedure that permits the connection design to be 

applied to framing sizes that are not identical to those used in the tests, while 

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retaining confidence that the connection will continue to behave as demonstrated 

by the testing. 

Testing is costly and time consuming, and these recommendations attempt to 

keep testing requirements as simple as possible. Test conditions should match the 

conditions in the structure as closely as possible, but it is recognized that test 

setups simultaneously account for the behavior and interdependence of many 

variables whose behavior is understood imprecisely. Where conditions in the 

structure differ significantly from the conditions implied in this section, additional 

testing to that recommended in these criteria may be required. 

3.9.1 Testing Procedures 

The testing program should follow the requirements of Appendix S of AISC Seismic with the 

exceptions and modifications discussed below. The program should include tests of at least two 

specimens for a given combination of beam and column size. The results of the tests should be 

capable of predicting the median value of the drift angle capacity for the performance states 

described in Table 3-13. The interstory drift angle q shall be defined as indicated in Figure 3-27. 

Acceptance criteria shall be as indicated in Section 3.9.2. 

Table 3-13 Interstory Drift Angle Limits for Various Performance Levels 

Performance Level Symbol Drift Angle Capacity 

Strength degradation qSD Taken as that value of q, from Figure 3-27 at which either failure of 

the connection occurs or the strength of the connection degrades to 

less than the nominal plastic capacity, whichever is less. 

Ultimate qU Taken as that value of q, from Figure 3-27 at which connection 

damage is so severe that continued ability to remain stable under 

gravity loading is uncertain. 

Column span mid−height to mid−height of story 

Undeformed centerline 

of beam 

LCL 

CL 

CL 

L 

D

q =

DCL 

Beam mid−span 

Figure 3-27 Angular Rotation of Test Assembly 

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q

q

1

2

3

4

5

6

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The following modifications and clarifications apply to Appendix S of the 1997 AISC 

Seismic Provisions as modified by Supplement No. 1: 

• In lieu of the requirements in Section S5.2, the size of the beam used in the test specimen 

shall be at least the largest depth and heaviest weight used in the structure. The column shall 

be selected to represent properly the anticipated inelastic action of the column in the real 

structure for the beam used in the test specimen. Extrapolation beyond the limits stated in this 

section is not recommended. 

• As an alternative to the loading sequence specified in Section S6.3, the FEMA/SAC loading 

protocol (Krawinkler et al., 2000) is considered acceptable. In the basic loading history, the 

cycles shall be symmetric in peak deformations. The history is divided into steps and the peak 

deformation of each step j is given as q j, a predetermined value of the drift angle. The 

loading history, shown in Table 3-14, is defined by the following parameters: 

q i = the peak deformation in load step j 

n j = the number of cycles to be performed in load step j 

Table 3-14 Numerical Values of qj and nj 

Load Step # Peak deformation q Number of cycles, n 

0.00375 6 

0.005 6 

0.0075 6 

0.01 4 

0.015 2 

0.02 2 

0.03 2 

Continue incrementing q in steps of 0.01 radians, and perform two 

cycles at each step until assembly failure occurs. Failure shall be 

deemed to occur when the peak loading in a cycle falls to 20% of that 

obtained at maximum load or, if the assembly has degraded, to a state 

at which stability under gravity load becomes uncertain. 

Commentary: The AISC Seismic Provisions (AISC, 1997) have been adopted by 

reference into the 1997 NEHRP Recommended Provisions for New Buildings. The 

AISC Seismic Provisions include, and require the use of, Appendix S – Qualifying 

Cyclic Tests of Beam-to-Column and Link-to-Column Connections, for 

qualification of connections that are not prequalified. Appendix S includes a 

complete commentary on the requirements. 

Under Appendix S, the Test Specimen must represent the largest beam 

anticipated in the project. The column must be selected to provide a flexural 

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strength consistent with the strong-column-weak-beam requirements and panel 

zone strength requirements. The permissive weight and size limits contained in 

Section S5.2 of Appendix S have been eliminated. 

The AISC loading history and acceptance criteria are described in terms of 

plastic rotation, while the FEMA/SAC loading protocol, acceptance criteria, and 

design recommendations, contained in these Recommended Criteria, are 

controlled by total drift angle, as previously defined. The engineer should ensure 

that appropriate adjustments are made when using the AISC loading history with 

these Recommended Criteria. 

The calculation of q illustrated in Figure 3-27 assumes that the top and the 

bottom of the column are restrained against lateral translation. The height of the 

test specimen column should be similar to that of the actual story height to 

prevent development of unrealistically large contributions to q from flexure of the 

column. In general, total drift angle is approximately equal to plastic rotation, 

measured as indicated in Figure 3-27, plus 0.01 radians. However, the engineer 

is cautioned that plastic rotation demand is often measured in different ways and 

may require transformation to be consistent with the measurement indicated in 

Figure 3-27. 

3.9.2 Acceptance Criteria 

For frames of typical configuration conforming in all respects to the applicable requirements 

of FEMA-302, and Chapter 2 of these Recommended Criteria, the median value of the interstory 

drift angle capacity at strength degradation, qSD, and at connection failure, qU, obtained from 

qualification testing shall not be less than indicated in Table 3-15. The coefficient of variation 

for these two parameters shall not exceed 10% unless the mean value, less one standard 

deviation, is also not less than the value indicated in Table 3-15. 

Table 3-15 Minimum Qualifying Total Interstory Drift Angle Capacities, qSD, and qU 

for OMF and SMF Systems 

Structural System Qualifying Drift Angle 

Capacity – Strength 

Degradation, qSD 

(radians) 

Qualifying Drift Angle 

Capacity – Ultimate, qU 

(radians) 

OMF 0.02 0.03 

SMF 0.04 0.06 

Note: 

Refer to Section 4.6.2.2.2 for definitions of qSD and qU 

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Where the clear-span-to-depth ratio of beams in the steel moment frame is less than 8, the 

qualifying total drift angle capacities indicated in Table 3-15 shall be increased to q'SD and q'U, 

given by equations 3-70 and 3-71, respectively: 

8d � L L - ¢� qS

¢

D = �1+ �qSD (3-70) L Ł L ł 

� L L - ¢� qU

¢ =+ �qU (3-71) �1

Ł L ł 

where: q'SD = Qualifying strength degradation drift angle capacity for spans with L/d < 8 

qSD = the basic qualifying strength degradation drift angle capacity, in accordance with 

Table 3-15 

q'U = the qualifying ultimate drift angle capacity, for spans with L/d < 8 

qU = the basic qualifying ultimate drift angle capacity, in accordance with Table 3-15 

L = the center-to-center spacing of columns, from Figure 3-1, inches. 

L' = the distance between points of plastic hinging in the beam, inches. 

d = depth of beam in inches 

Commentary: This section sets criteria for use in project-specific qualification of 

connections, in accordance with Section 3.9, and for development of new 

connection prequalifications in accordance with Section 3.10 of these 

Recommended Criteria. Two interstory drift angle capacities are addressed. The 

values indicated in Table 3-15 formed the basis for extensive probabilistic 

evaluations of the performance capability of various structural systems, reported 

in FEMA-355F, State of the Art Report on Performance Prediction and 

Evaluation. These probabilistic evaluations indicate a high confidence, on the 

order of 90%, that regular, well-configured frames meeting the requirements of 

FEMA-302 and constructed with connections having these capabilities, can meet 

the intended performance objectives with regard to protection against global 

collapse, and moderate confidence, on the order of 50%, that connections can 

resist maximum considered earthquake demands without local life-threatening 

damage. 

Connection details with capacities lower than those indicated in this section 

should not be incorporated in structures unless a specific probabilistic analysis 

using the performance evaluation procedures contained in Chapter 4 and 

Appendix A of these Recommended Criteria indicates that an acceptable level of 

confidence of adequate performance can be obtained. 

Connections in frames where beam span-to-depth ratios are less than those 

used for the prequalification testing, will experience larger flange strains, at the 

plastic hinges at a particular frame drift, than those tested. For this reason, 

connections used in such frames need to be qualified for larger drifts as indicated 

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by the formulae in this section, unless the frames are designed to experience 

proportionately lower drifts than permitted by FEMA-302. 

3.9.3 Analytical Prediction of Behavior 

Connection qualification should include development of an analytical procedure to predict 

the limit states of the connection assembly, as demonstrated by the qualification tests. The 

analytical procedure should permit identification of the strength and deformation demands and 

limit states on various elements of the assembly at the various stages of behavior. The analytical 

procedure should be sufficiently detailed to permit design of connections employing members 

similar to those tested within the limits identified in Section S5.2 of AISC Seismic. 

Commentary: It is important for the designer to have an understanding of the 

limiting behaviors of any connection detail so that the detail may be designed and 

specified on a rational basis for assemblies that vary, within specified limits, from 

those tested. 

3.10 Prequalification Testing Criteria 

This section provides guidelines for prequalification of connections for which there is no 

current prequalification or to extend the parametric limitations for prequalification listed in 

Sections 3.5 and 3.6. Prequalification includes a program of connection assembly prototype 

testing supplemented by a suitable analytical procedure that permits prediction of behaviors 

identified in the testing program. 

Commentary: The purpose of this section is to provide recommended procedures 

for prequalification of a connection that is not currently prequalified in these 

Recommended Criteria or to extend the range of member sizes that may be used 

with currently prequalified connections for general application. These criteria are 

intended to require significantly more testing than are required for a projectspecific qualification program, as once a connection is prequalified, it can see 

wide application. Prequalification of a connection should incorporate the testing 

described in this section as well as due consideration of the four criteria 

described in the Commentary for Section 3.4. 

The potential for limit states leading to local collapse (i.e. loss of gravity-load 

capacity) is an important consideration in evaluating the performance of a 

prototype connection. Establishing this limit state required by Section 3.9.1 will 

necessitate imposing large deformations on the connection. This will require 

loading setups capable of delivering long strokes while withstanding 

correspondingly large out-of-plane or large torsional deformations. Many tests 

are terminated before the ultimate failure of the connection to protect the loading 

apparatus. These early terminations will limit the range over which a connection 

may be prequalified. 

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3.10.1 Prequalification Testing 

Testing and acceptance criteria should follow the recommendations in Section 3.9 except that 

at least five non-identical test specimens shall be used. The resulting range of member sizes that 

will be prequalified should be limited to the range represented by the tested specimens. 

3.10.2 Extending the Limits on Prequalified Connections 

Testing and acceptance criteria should follow the recommendations in Section 3.9 except that 

at least two non-identical test specimens shall be tested. The resulting range of member size that 

will be prequalified should be limited to the those contained in the data base of tests for the 

connection type. 

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4. Performance Evaluation 

4.1 Scope 

This chapter provides simplified criteria for evaluating the probable seismic performance of 

welded steel moment-frame buildings. It may, as an option, be used in parallel with the design 

procedures of Chapters 2 and 3 of these Recommended Criteria to design steel moment-frame 

buildings for alternative performance capabilities and also to quantify the ability of a specific 

design to achieve desired performance objectives. It includes definitions of performance 

objectives, discussions of expected performance of buildings conforming to FEMA-302, NEHRP 

Recommended Provisions for Seismic Regulations for New Buildings and Other Structures 

(BSSC, 1997a), and procedures for estimating a level of confidence that a building will be able to 

provide a desired level of performance for specified earthquake hazards. It is applicable only to 

well-configured, regular structures as defined in FEMA-302. A more detailed procedure, 

applicable to irregular structures and performance objectives based on deterministic earthquake 

scenarios is presented in Appendix A of these Recommended Criteria. 

Commentary: These criteria only address methods of evaluating structural 

performance of welded steel moment-frame buildings. Although the performance 

of nonstructural components of buildings is critically important to the way in 

which buildings are used following an earthquake, treatment of this topic is 

beyond the scope of this document. FEMA-273, NEHRP Guidelines for the 

Seismic Rehabilitation of Buildings, provides a more complete set of 

recommendations with regard to evaluating the performance of nonstructural 

components. 

FEMA-355F, State of the Art Report on Performance Prediction and 

Evaluation, presents in detail the basis for the procedures contained herein and 

the derivation of the various parameters used in the procedures. 

4.2 Performance Definition 

The performance evaluation procedures contained in these Recommended Criteria permit 

estimation of a level of confidence that a structure will be able to achieve a desired performance 

objective. Each performance objective consists of the specification of a structural performance 

level and a corresponding hazard level, for which that performance level is to be achieved. For 

example, a design may be determined to provide a 95% level of confidence that the structure will 

provide Collapse Prevention or better performance for earthquake hazards with a 2% probability 

of exceedance in 50 years, or a 50% level of confidence that the structure will provide Immediate 

Occupancy or better performance, for earthquake hazards with a 50% probability of exceedance 

in 50 years. 

Commentary: The performance evaluation procedures contained in these 

Recommended Criteria are based on an approach first developed in FEMA-273. 

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Chapter 4: Performance Evaluation Moment-Frame Buildings 

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However, substantial modifications have been made to the procedures presented 

in that document. 

In FEMA-273, performance objectives are expressed in a deterministic 

manner. Each performance objective consists of the specification of a limiting 

damage state, termed a performance level, together with a specification of the 

ground motion intensity for which that (or better) performance is to be provided. 

This implies a warranty that if the specified ground motion is actually 

experienced by a building designed using the FEMA-273 procedures, damage will 

be no worse than that indicated in the performance objective. 

In reality, it is very difficult to predict with certainty how much damage a 

building will experience for a given level of ground motion. This is because there 

are many factors that affect the behavior and response of a building (such as the 

stiffness of nonstructural elements, the strength of individual building 

components, and the quality of construction) that cannot be precisely defined, and 

also because the analysis procedures used to predict building response are not 

completely accurate. In addition, the exact character of the ground motion that 

will actually affect a building is itself uncertain. Given these uncertainties, it is 

inappropriate to imply that performance can be predicted in an absolute sense, 

and correspondingly, that it is absolutely possible to produce designs that will 

achieve desired performance objectives. 

In recognition of this, these Recommended Criteria adopt a reliability-based 

probabilistic approach to performance evaluation that explicitly acknowledges 

these inherent uncertainties. These uncertainties are expressed in terms of a 

confidence level. If an evaluation indicates a high level of confidence, for 

example 90 or 95% that a performance objective can be achieved, then this means 

it is very likely (but not guaranteed) that the building will be capable of meeting 

the desired performance. If lower confidence is calculated, for example 50%, this 

is an indication that the building may not be capable of meeting the desired 

performance objective. If still lower confidence is calculated, for example 30%, 

then this indicates the building will likely not be able to meet the desired 

performance objective. Increased confidence in a building’s ability to provide 

specific performance can be obtained in three basic ways. 

• Providing the building with greater earthquake resistance, for example, by designing 

the structure to be stiffer and stronger. 

• Reducing some of the uncertainty inherent in the performance evaluation process 

through the use of more accurate structural models and analyses and better data on 

the building’s configuration, strength and stiffness. 

• More accurately characterizing the uncertainties inherent in the performance 

evaluation process, for example, by using the more exact procedures of Appendix A of 

these Recommended Criteria. 

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Refer also to the commentary in Section 4.2.1.2 for additional discussion of the 

probabilistic approach adopted by this document. 

4.2.1 Hazard Specification 

4.2.1.1 General 

Earthquake hazards include direct ground fault rupture, ground shaking, liquefaction, lateral 

spreading, and land sliding. Of these various potential hazards, the one that effects the largest 

number of structures and causes the most widespread damage is ground shaking. Ground 

shaking is the only earthquake hazard that the structural design provisions of the building codes 

directly address. However, for structures located on sites where any of the other hazards can 

result in significant ground deformation, these hazards should also be considered in a structural 

performance evaluation. 

4.2.1.2 Ground Shaking 

Ground shaking hazards are typically characterized by a hazard curve, which indicates the 

probability that a given value of a ground motion parameter, for example peak ground 

acceleration, will be exceeded over a certain period of time, and by acceleration response spectra 

or ground motion accelerograms that are compatible with the values of the ground motion 

parameters obtained from the hazard curve and the local site geology. The ground shaking 

hazard maps contained in FEMA-302 and provided with FEMA-273 have been prepared based on 

hazard curves that have been developed by the United States Geologic Survey for a grid-work of 

sites encompassing the United States and its territories. FEMA-302 defines two specific levels of 

hazard for consideration in design and specifies methods for developing response spectra for 

each of these levels. The two levels are: 

1. Maximum Considered Earthquake (MCE) ground shaking. This is the most severe level of 

ground shaking that is deemed appropriate for consideration in the design process for 

building structures, though not necessarily the most severe level of ground shaking that could 

ever be experienced at a site. In most regions, this ground shaking has a 2% probability of 

exceedance in 50 years, or roughly a 2,500 year mean recurrence interval. In regions of very 

high seismicity, near major active faults, the MCE ground shaking level is limited by a 

conservative, deterministic estimate of the ground shaking resulting from a maximum 

magnitude earthquake on the known active faults in the region. The probability that such 

deterministic ground shaking will be experienced at a site can vary considerably, depending 

on the activity rate of the individual fault. Refer to FEMA-303, Commentary to the NEHRP 

Recommended Provisions for Seismic Regulations of New Buildings and Other Structures, 

for more detailed information on this issue. 

2. Design Earthquake (DE) ground shaking. This is the ground shaking level upon which 

design lateral forces, used as the basis for analysis and design in FEMA-302, are based. It is 

defined as a spectrum that is 2/3 of the shaking intensity calculated for the MCE spectrum, at 

each period. The probability that DE ground shaking will be experienced varies, depending 

on the regional and in some cases, site seismicity. 

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Chapter 4: Performance Evaluation Moment-Frame Buildings 

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Commentary: The mean recurrence interval for Design Earthquake (DE) ground 

shaking will vary depending on regional and site seismicity. In areas of low 

seismicity the DE return period will generally range between 750-1,250 years and 

will remain relatively constant across neighboring communities. In areas of high 

seismicity the recurrence interval may range between 300-600 years and can vary 

significantly within a distance of a few miles. 

Performance evaluation conducted in accordance with these guidelines may be performed for 

any level of ground shaking. Ground shaking will typically be determined probabilistically, i.e., 

based on the probability that shaking of the specified intensity will be experienced at a site. 

Ground shaking must be characterized by an acceleration response spectrum or a suite of ground 

motion accelerograms compatible with that spectrum. In addition, a coefficient k that relates the 

rate of change in ground motion intensity with change in probability, is required. FEMA-273 

provides guidelines for development of ground motion response spectra at different probabilities 

of exceedance. The procedures of this chapter use a default value for the coefficient, k, as 

described in the commentary. Performance evaluation for deterministic ground motion based on 

specific earthquake scenarios, for example an earthquake of given magnitude on a specific fault, 

can also be performed. Appendix A provides procedures that may be used for deterministically 

defined hazards. 

Commentary: Detailed guidelines on ground motion estimation and 

characterization are beyond the scope of this publication. Those interested in 

such information are referred to FEMA-303 and FEMA-274, Commentary to the 

NEHRP Guidelines for the Rehabilitation of Buildings, and references noted 

therein. 

Although Section 4.2 of these Recommended Criteria indicates that 

performance objectives are an expression of the desired performance for a 

building, given that ground motion of certain intensity is experienced, this is a 

significant simplification. In reality, the performance objectives are statements of 

the total probability that damage experienced by a building in a period of years 

will be more severe than the desired amount (performance level), given our 

knowledge of the site seismicity. Although it is transparent to the user, this is 

obtained by integrating the conditional probability that building response exceeds 

the limiting response for a performance level, given a ground motion intensity, 

over the probability of experiencing different intensities of ground motion, as 

represented by the site hazard curve, and specifically, the coefficient k, which is 

the logarithmic slope of the hazard curve, at the desired hazard level. Thus, a 

performance objective that is stated as “meeting collapse prevention performance 

for ground motion with a 2% probability of exceedance in 50 years” should more 

correctly be stated as being “less than a 2% chance in 50 years that damage more 

severe than the collapse prevention level will occur, given the mean definition of 

seismicity.” 

It is important to note that the procedures contained in this chapter neglect 

uncertainties associated with the definition of the seismicity, that is, the intensity 

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Moment-Frame Buildings Chapter 4: Performance Evaluation

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of ground shaking at various probabilities. Such uncertainties can be as large, 

and perhaps larger, than the uncertainties associated with structural performance 

estimation. Thus, the confidence calculated in accordance with the procedures of 

this chapter is really a confidence associated with structural performance, given 

the presumed seismicity. 

The simplified procedures presented in this chapter have been developed 

using hazard parameters typical of coastal California. They can be 

conservatively applied in regions of lower seismicity without the need to 

determine site-specific hazard parameters. However, accurate definition of the 

hazard is a critical part of the performance evaluation procedures contained 

herein and in regions of lower seismicity, may result in calculation of higher 

confidence. Appendix A of these Recommended Criteria presents more detailed 

procedures that may be used to consider directly the site-specific characteristics 

of hazard in the evaluation of performance. 

4.2.1.3 Other Hazards 

In order to predict reliably the probable performance of a structural design, it is necessary to 

determine if earthquake hazards other than ground shaking, including direct ground fault rupture, 

liquefaction, lateral spreading, and land sliding are likely to occur at a site and to estimate the 

severity of these effects. The severity of ground fault rupture, lateral spreading and land sliding 

is characterized by an estimate of permanent ground deformation. The severity of liquefaction is 

characterized by an estimate of the potential loss in bearing strength of subsoil layers and 

permanent ground settlement. In order to determine the performance of a structure that is subject 

to these hazards, the effects of the projected ground displacements should be evaluated using a 

mathematical model of the structure. The severity of these hazards (i.e. probability of 

exceedance) used in performance evaluation should be compatible with that used in the 

specification of ground shaking hazards. 

Commentary: Most sites are not at significant risk from earthquake hazards 

other than ground shaking. However, these hazards can be very destructive to 

structures located on sites where they occur. Accurate determination of the 

propensity of a site to experience these hazards requires site-specific study by a 

competent earth scientist or geotechnical engineer. Guidelines on such 

assessments are beyond the scope of these Recommended Criteria. 

4.2.2 Performance Levels 

Building performance is a combination of the performance of both structural and 

nonstructural components. Table 4-1 describes the overall levels of structural and nonstructural 

damage that may be expected of buildings meeting two performance levels, termed Collapse 

Prevention and Immediate Occupancy. These performance descriptions are not precise and 

variation among buildings must be expected, within the same Performance Level. The structural 

performance levels are presented in Section 4.2.2.2. 

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Chapter 4: Performance Evaluation Moment-Frame Buildings 

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Table 4-1 Building Performance Levels 

Building Performance Levels 

Collapse Prevention Level Immediate Occupancy Level 

Overall Damage Severe Light 

General Little residual stiffness and strength, but 

gravity loads are supported. Large 

permanent drifts. Some exits may be 

blocked. Exterior cladding may be 

extensively damaged and some local 

failures may occur. Building is near 

collapse. 

Structure substantially retains original 

strength and stiffness. Minor cracking 

of facades, partitions, ceilings, and 

structural elements. Elevators can be 

restarted. Fire protection operable. 

Nonstructural 

components 

Extensive damage. Equipment and contents are generally 

secure, but may not operate due to 

mechanical failure or lack of utilities. 

Comparison with 

performance intended 

by FEMA-302 for 

SUG1

-I buildings 

when subjected to the 

Design Earthquake 

Significantly more damage and greater 

risk. 

Much less damage and lower risk. 

Comparison with 

performance intended 

by FEMA-302 for 

SUG1

-I buildings 

when subjected to the 

Maximum Considered 

Earthquake 

Same level of performance Much less damage and lower risk. 

Note: 1. SUG = Seismic Use Group 

Commentary: Building performance is expressed in terms of building 

performance levels. These building performance levels are discrete damage 

states selected from among the infinite spectrum of possible damage states that 

steel moment-frame buildings could experience as a result of earthquake 

response. The particular damage states identified as building performance levels 

have been selected because these performance levels have readily identifiable 

consequences associated with the postearthquake disposition of the building that 

are meaningful to the building user community and also because they are 

quantifiable in technical terms. These include the ability to resume normal 

functions within the building, the advisability of postearthquake occupancy, and 

the risk to life safety. 

Although a building’s performance is a function of the performance of both 

structural systems and nonstructural components and contents, only the structural 

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performance levels are defined in these Recommended Criteria. The reference to 

nonstructural components above is to remind the reader of the probable 

performance of these elements at the various performance levels. 

4.2.2.1 Nonstructural Performance Levels 

These Recommended Criteria only address methods of evaluating structural performance of 

steel moment-frame buildings. Although the performance of nonstructural components of 

buildings are critically important to the way in which buildings are used following an earthquake, 

treatment of this topic is beyond the scope of this document. FEMA-273 provides a more 

complete set of recommendations for evaluating the performance of nonstructural components. 

4.2.2.2 Structural Performance Levels 

Two discrete structural performance levels, Collapse Prevention and Immediate Occupancy, 

are defined in these Recommended Criteria. Table 4-2 relates these structural performance levels 

to the limiting damage states for common framing elements of steel moment-frame buildings. 

Acceptance criteria, which relate to the permissible interstory drifts and earthquake-induced 

forces for the various elements of steel moment-frame buildings, are tied directly to these 

structural performance levels and are presented in later sections of these Recommended Criteria. 

Commentary: FEMA-273 defines three structural performance levels, Immediate 

Occupancy, Life Safety, and Collapse Prevention, and also defines two 

performance ranges. These performance ranges, rather than representing 

discrete damage states, span the entire spectrum of potential damage states 

between no damage and total damage. No acceptance criteria are provided for 

these performance ranges in FEMA-273. Rather, these must be determined on a 

project-specific basis, by interpolation or extrapolation from the criteria provided 

for the three performance levels. Performance ranges, as such, are not defined in 

these Recommended Criteria. However, compatible with the FEMA-273 

approach, users have the ability to create their own, custom performance levels, 

and to develop acceptance criteria for these levels, based on interpolation 

between the two performance levels, to suit the needs of a specific project. When 

such interpolation is performed, it is not possible to associate a confidence level 

with achievement of these intermediate performance definitions. 

The Life Safety performance level contained in FEMA-273 and FEMA-302 is 

not included in these Recommended Criteria. As defined in FEMA-273 and 

FEMA-302, the Life Safety level is a damage state in which significant damage 

has been sustained, although some margin remains against either partial or total 

collapse. In FEMA-273 this margin is taken as 1/3. That is, it is anticipated that 

a ground motion level that is 1/3 larger than that which results in the Life Safety 

performance level for a building would be required to bring the building to the 

Collapse Prevention level. In FEMA-302, this margin is taken as 1⁄2, i.e., it is 

believed that buildings designed for Life Safety performance can experience 

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approximately 50% greater motion before they reach the Collapse Prevention 

level. Due to the somewhat arbitrary definition of this performance level, and the 

fact that different guidelines and codes have selected alternative definitions for it 

(as described above), the Life Safety level has not been included in these 

Recommended Criteria. However, as with the performance ranges, users desiring 

to evaluate buildings for the Life Safety performance level may do so by 

interpolating between the acceptance criteria provided for the Collapse 

Prevention and Immediate Occupancy levels. 

Table 4-2 Structural Performance Levels 

Structural Performance Levels 

Elements Collapse Prevention Immediate Occupancy 

Girder Extensive distortion; local yielding 

and buckling. A few girders may 

experience partial fractures 

Minor local yielding and buckling at 

a few places. 

Column Moderate distortion; some columns 

experience yielding. Some local 

buckling of flanges 

No observable damage or distortion 

Beam-Column 

Connections 

Many fractures with some 

connections experiencing near total 

loss of capacity 

Less than 10% of connections 

fractured on any one floor; minor 

yielding at other connections 

Panel Zone Extensive distortion Minor distortion 

Column Splice No fractures No yielding 

Base Plate Extensive yielding of anchor bolts 

and base plate 

No observable damage or distortion 

Interstory Drift Large permanent Less than 1% permanent 

4.2.2.2.1 Collapse Prevention Performance Level 

The Collapse Prevention structural performance level is defined as the postearthquake 

damage state in which the structure is on the verge of experiencing partial or total collapse. 

Substantial damage to the structure has occurred, potentially including significant degradation in 

the stiffness and strength of the lateral-force-resisting system, large permanent lateral 

deformation of the structure, and, to a more limited extent, degradation in the vertical-loadcarrying capacity. However, all significant components of the gravity-load-resisting system must 

continue to carry their gravity-load demands. The structure may not be technically or 

economically practical to repair and is not safe for re-occupancy; aftershock activity could 

credibly induce collapse. 

4.2.2.2.2 Immediate Occupancy Performance Level 

The Immediate Occupancy structural performance level is defined as the postearthquake 

damage state in which only limited structural damage has occurred. Damage is anticipated to be 

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so slight that it would not be necessary to inspect the building for damage following the 

earthquake, and such little damage as may be present would not require repair. The basic 

vertical- and lateral-force-resisting systems of the building retain nearly all of their preearthquake strength and stiffness. The risk of life-threatening injury as a result of structural 

damage is very low. Buildings meeting this performance level should be safe for immediate 

postearthquake occupancy, presuming that damage to nonstructural components is suitably light 

and that needed utility services are available. 

Commentary: When a building is subjected to earthquake ground motion, a 

pattern of lateral deformations that varies with time is induced in the structure. 

At any given point in time, a particular state of lateral deformation will exist in 

the structure, and at some time within the period in which the structure is 

responding to the ground motion, a maximum pattern of deformation will occur. 

At relatively low levels of ground motion, the deformations induced within the 

building will be limited, and the resulting stresses that develop within the 

structural components will be within their elastic range of behavior. Within this 

elastic range, the structure will experience no damage. All structural components 

will retain their original strength, stiffness and appearance, and when the ground 

motion stops, the structure will return to its pre-earthquake condition. 

At more severe levels of ground motion, the lateral deformations induced in 

the structure will be larger. As these deformations increase, so will demands on 

the individual structural components. At different levels of deformation, 

corresponding to different levels of ground motion severity, individual 

components of the structure will be strained beyond their elastic range. As this 

occurs, the structure starts to experience damage in the form of buckling, yielding 

and fracturing of the various components. As components become damaged, they 

degrade in stiffness, and some elements will begin to lose their strength. In 

general, when a structure has responded to ground motion within this range of 

behavior, it will not return to its pre-earthquake condition when the ground 

motion stops. Some permanent deformation may remain within the structure and 

damage may be evident throughout. Depending on how far the structure has been 

deformed, and in what pattern, the structure may have lost a significant amount of 

its original stiffness and, possibly, strength. 

Brittle elements are not able to sustain inelastic deformations and will fail 

suddenly; the consequences may range from local and repairable damage to 

collapse of the structural system. At higher levels of ground motion, the lateral 

deformations induced in a structure will strain a number of elements to a point at 

which the elements degrade in stiffness and strength, or as a result of P-D effects, 

the structure loses stability. Eventually, partial or total collapse of the structure 

can occur. The structural performance levels relate the extent of a building’s 

response to earthquake hazards to these various possible damage states. 

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At the Immediate Occupancy Level, degradation of strength and stiffness in 

beam-column connections is limited to approximately 10% of the connections on 

any given floor and throughout the structure as a whole. The structure retains a 

significant portion of its original stiffness and most, if not all, of its strength, 

although some slight permanent drift may result. At the Collapse Prevention 

level, the building has experienced extreme damage. If laterally deformed beyond 

this point, the structure can experience instability and can collapse. 

4.3 Evaluation Approach 

The basic process of performance evaluation, as contained in these Recommended Criteria is 

to develop a mathematical model of the structure and to evaluate its response to the earthquake 

hazards by one or more methods of structural analysis. The structural analysis is used to predict 

the value of various structural response parameters. These include: 

• interstory drift 

• axial forces on individual columns 

These structural response parameters are related to the amount of damage experienced by 

individual structural components as well as the structure as a whole. For each performance level, 

these Recommended Criteria specify acceptance criteria (median estimates of capacity) for each 

of the design parameters indicated above. Acceptability of structural performance is evaluated 

considering both local performance (element level) and global performance. Acceptance criteria 

have been developed on a reliability basis, incorporating demand and resistance factors related to 

the uncertainty inherent in the evaluation process and variation inherent in structural response, 

such that a confidence level can be established with regard to the ability of a structure to actually 

provide specific performance at selected probabilities of exceedance. 

Once an analysis is performed, predicted demands are adjusted by two factors, an analytical 

uncertainty factor ga, which corrects the analytically predicted demands for bias and uncertainty 

inherent in the analytical technique, and demand variability factor g, which accounts for other 

sources of variability in structural response. These predicted demands are compared against 

acceptance criteria, which have been modified by resistance factors f to account for uncertainties 

and variation inherent in structural capacity. Procedures are given to calculate the level of 

confidence provided by a design to achieve specific performance objectives, based on the ratio of 

factored demand to factored capacity. If the predicted level of confidence is inadequate, then 

either more detailed analyses should be performed to improve the level of confidence attained 

with regard to performance, through the attainment of better understanding of the structure’s 

probable behavior and modification of the demand and capacity factors, or the design can be 

revised such that a sufficient level of confidence can be attained given the level of understanding. 

If it is deemed appropriate to upgrade a design to improve its probable performance, an iterative 

approach consisting of trial design, followed by verification analysis, evaluation of design 

parameters against acceptance criteria, and calculation of confidence level is repeated until an 

acceptable design solution is found. Procedures for estimating confidence are found in Section 

4.6. 

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Commentary: These Recommended Criteria adopt a demand and resistance 

factor design (DRFD) model for performance evaluation. This approach is 

similar to the Load and Resistance Factor design (LRFD) approach adopted by 

AISC LRFD except that the LRFD provisions are conducted on an element basis, 

rather than a structural system basis, and demands in this document can be drifts 

as well as forces and stresses. The purpose of this DRFD approach is to allow 

characterization of the confidence level inherent in a design to a specific 

performance objective. 

The factored interstory drift demand gagD calculated from the analysis, 

represents a median estimate of the probable maximum interstory drift demand, at 

the desired hazard level. Tables in these Recommended Criteria provide 

interstory drift capacities for the two performance levels for regular, wellconfigured structures, dependent on structural system and connection type, as 

well as resistance factors f that adjust the estimated capacity of the structure to 

median values. Appendix A provides procedures for determination of f factors 

for connections for which project-specific qualification testing is performed and a 

procedure that may be used to determine interstory drift capacities for irregular 

structures. 

Once the factored demands and capacities are determined, an index 

parameter l is calculated from the ratio of the factored demands and capacities, 

as indicated in Section 4.6. The value of l is then used to determine an 

associated confidence level based on tabulated values related to the uncertainty 

inherent in the estimation of the building’s demands and capacities. 

4.4 Analysis 

In order to evaluate the performance of a steel moment-frame building it is necessary to 

construct a mathematical model of the structure that represents its strength and deformation 

characteristics, and to conduct an analysis to predict the values of various design parameters 

when it is subjected to design ground motion. This section provides guidelines for selecting an 

appropriate analysis procedure and for modeling. General requirements for the mathematical 

model are presented in Section 4.5. 

4.4.1 Alternative Procedures 

Four alternative analytical procedures are available for use in performance evaluation of steel 

moment-frame buildings. The basic analytical procedures are described in detail in FEMA-273. 

This section provides supplementary guidelines on the applicability of the FEMA-273 procedures 

and also provides supplemental modeling recommendations. The four procedures are: 

• linear static procedure – an equivalent lateral force technique, similar, but not identical, to 

that contained in many model building code provisions, 

• linear dynamic procedure – an elastic, modal, response-spectrum analysis, 

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• nonlinear static procedure – a simplified nonlinear analysis procedure in which the forces and 

deformations induced by a monotonically increasing lateral loading is evaluated using a 

series of incremental elastic analyses of structural models that are sequentially degraded to 

represent the effects of structural nonlinearity, 

• nonlinear dynamic procedure – a nonlinear dynamic analysis procedure in which the response 

of a structure to a suite of ground motion histories is determined through numerical 

integration of the equations of motion for the structure. Structural stiffness is altered during 

the analysis to conform to nonlinear hysteretic models of the structural components. 

Commentary: The purpose of the structural analyses performed as part of the 

performance evaluation process is to predict the values of key response 

parameters that are indicative of the structure’s performance when it is subjected 

to ground motion. Once the values of these response parameters are predicted, 

the structure is evaluated for adequacy (appropriate level of confidence of 

achieving desired performance) using the basic approach outlined in Section 4.6. 

Analyses performed in support of design, as required by FEMA-302, evaluate 

the strength and deformation of the structure when it is subjected to a somewhat 

arbitrary level of loading. The loading is based on, but substantially reduced 

from, that predicted by an elastic analysis of the structure’s dynamic response to 

the expected ground motions. Specifically, the loading is reduced by a factor R 

to account approximately for the beneficial effects of inelastic response. 

Analyses conducted in support of performance evaluation under these 

Recommended Criteria take a markedly different approach. Rather than 

evaluating the forces and deformations induced in the structure under arbitrarily 

reduced loading levels, these analysis procedures attempt to predict, within 

probabilistically defined bounds, the actual values of the important response 

parameters in response to design ground motion. 

The ability of the performance evaluation to estimate reliably the probable 

performance of the structure is dependent on the ability of the analysis procedure 

to predict the values of these response parameters within acceptable levels of 

confidence. The linear dynamic procedure is able to provide relatively reliable 

estimates of the response parameters for structures that exhibit elastic, or near 

elastic, behavior. The linear static procedure inherently has more uncertainty 

associated with its estimates of the response parameters because it accounts less 

accurately for the dynamic characteristics of the structure. The nonlinear static 

procedure is more reliable than the linear procedures in predicting response 

parameters for structures that exhibit significant nonlinear behavior, particularly 

if they are irregular. However, it does not accurately account for the effects of 

higher mode response. If appropriate modeling is performed, the nonlinear 

dynamic approach is most capable of capturing the probable behavior of the real 

structure in response to ground motion. However, there are considerable 

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uncertainties associated even with the values of the response parameters 

predicted by this technique. 

4.4.2 Procedure Selection 

Table 4-3 indicates the recommended analysis procedures for various performance levels and 

conditions of structural regularity. 

4.4.3 Linear Static Procedure 

4.4.3.1 Basis of the Procedure 

Linear static procedure (LSP) analysis of steel moment-frame structures should be conducted 

in accordance with the recommendations of FEMA-273, except as noted herein. In this 

procedure, lateral forces are applied to the masses of the structure, and deflections and 

component forces under this applied loading are determined. Calculated internal forces typically 

will exceed those that the building can develop, because anticipated inelastic response of 

components and elements is not directly recognized by the procedure. The predicted interstory 

drifts and column axial forces are evaluated using the procedures of Section 4.6. 

Commentary: The linear static procedure is a method of estimating the response 

of the structure to earthquake ground shaking by representing the effects of this 

response through the application of a series of static lateral forces applied to an 

elastic mathematical model of the structure and its stiffness. The forces are 

applied to the structure in a pattern that represents the typical distribution of 

inertial forces in a regular structure responding in a linear manner to the ground 

shaking excitation, factored to account, in an approximate manner, for the 

probable inelastic behavior of the structure. It is assumed that the structure’s 

response is dominated by the fundamental mode and that the lateral drifts induced 

in the elastic structural model by these forces represent a reasonable estimate of 

the actual deformation of the building when responding inelastically. 

In the LSP, the building is modeled with linearly-elastic stiffness and 

equivalent viscous damping that approximate values expected for loading to near 

the yield point. The static lateral forces, whose sum is equal to the pseudo lateral 

load, represent earthquake demands for the LSP. The magnitude of the pseudo 

lateral load has been selected with the intention that when it is applied to the 

linearly elastic model of the building it will result in design displacement 

amplitudes approximating maximum displacements that are expected during the 

design earthquake. If the building responds essentially elastically to the design 

earthquake, the calculated internal forces will be reasonable approximations of 

those expected during the design earthquake. If the building responds 

inelastically to the design earthquake, as will commonly be the case, the internal 

forces that would develop in the yielding building will be less than the internal 

forces calculated on an elastic basis, but the predicted interstory drifts will 

approximate those that would actually occur in the structure. 

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Table 4-3 Analysis Procedure Selection Criteria 

Structural Characteristics Analytical Procedure 

Performance 

Level 

Fundamental 

Period, T 

Regularity Ratio of Column to 

Beam Strength 

Linear 

Static 

Linear 

Dynamic 

Nonlinear 

Static 

Nonlinear 

Dynamic 

Immediate 

Occupancy 

T < 3.5Ts 

1 Regular or 

Irregular 

Any Condition Permitted Permitted Permitted Permitted 

T > 3.5 Ts 

1 Regular or 

Irregular 

Any Conditions Not 

Permitted 

Permitted Not 

Permitted 

Permitted 

Collapse 

Prevention 

T < 3.5Ts 

1 Regular2 Strong Column3 Permitted Permitted Permitted Permitted 

Weak Column3 Not 

Permitted 

Not 

Permitted 

Permitted Permitted 

Irregular2 Any Conditions Not 

Permitted 

Not 

Permitted 

Permitted Permitted 

T > 3.5Ts Regular Strong Column3 Not 

Permitted 

Permitted Not 

Permitted 

Permitted 

Weak Column3 Not 

Permitted 

Not 

Permitted 

Not 

Permitted 

Permitted 

Irregular2 Any Conditions Not 

Permitted 

Not 

Permitted 

Not 

Permitted 

Permitted 

Notes: 

1. Ts is the period at which the response spectrum transitions from a domain of constant response acceleration 

(the plateau of the response spectrum curve) to one of constant spectral velocity. Refer to FEMA-273 or 

FEMA-302 for more information. 

2. Conditions of regularity are as defined in FEMA-273. These conditions are significantly different from those 

defined in FEMA-302. 

3. A structure qualifies as having a strong column condition if at every floor level, the quantity SMprc /SMprb is 

greater than 1.0, where SMprc and SMprb are the sum of the expected plastic moment strengths of the columns 

and beams, respectively, that participate in the moment-resisting framing in a given direction of structural 

response. 

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The performance of steel moment-frame structures is most closely related to 

the total inelastic deformation demands on the various elements that comprise the 

structure, such as plastic rotation demands on beam-column assemblies and 

tensile demands on column splices. Linear analysis methods do not permit direct 

evaluation of such demands. However, through a series of analytical evaluations 

of typical buildings for a number of earthquake records, it has been possible to 

develop statistical correlation between the interstory drift demands predicted by a 

linear analysis and the actual inelastic deformation demands determined by more 

accurate nonlinear methods. These correlation relationships are reasonably 

valid for regular buildings, using the definitions of regularity in FEMA-273. 

Although performance of steel moment-frame structures is closely related to 

interstory drift demand, there are some failure mechanisms, notably, the failure of 

column splices, that are more closely related to strength demand. However, since 

inelastic structural behavior affects the strength demand on such elements, linear 

analysis is not capable of directly predicting these demands, except when the 

structural response is essentially elastic. Therefore, when linear static analysis is 

performed for structures that respond in an inelastic manner, column axial 

demands should be estimated using a supplementary plastic analysis approach. 

The linear static procedure is based on the assumption that the distribution of 

deformations predicted by an elastic analysis is similar to that which will occur in 

actual nonlinear response. This assumption is inaccurate and can become more 

so for structures that are highly irregular, that have response dominated by 

higher modes, or that experience large inelastic demands. It is for these reasons 

that alternative methods of analysis are recommended for irregular structures 

and structures with relatively long fundamental periods of vibration. 

4.4.3.2 Period Determination 

A fundamental period shall be calculated for each of two orthogonal directions of building 

response, by one of the following three methods. 

Method 1. Eigenvalue (dynamic) analysis of the mathematical model of the building. The 

model for buildings with flexible diaphragms shall consider representation of diaphragm 

flexibility unless it can be shown that the effects of omission will not be significant. 

Method 2. Evaluation of the following equation: 

T = Ch 0.8 (4-1) t n

where 

T = fundamental period (in seconds) in the direction under consideration, 

Ct = 0.028 for steel moment frames, 

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hn = height (in feet) to the roof level, above the base. 

Method 3. The fundamental period of a one-story building with a single-span, flexible 

diaphragm may be calculated as: 

Dd 

0.5 T = (0.1D (4-2) w + 0.078) 

where Dw and Dd are in-plane frame and diaphragm displacements, respectively, in inches, 

due to a lateral load, in the direction under consideration, equal to the weight tributary to the 

diaphragm. For multiple-span diaphragms, a lateral load equal to the gravity weight tributary 

to the diaphragm span under consideration should be applied to each diaphragm span to 

calculate a separate period for each diaphragm span. The loads from each diaphragm should 

then be distributed to the frames using tributary load assumptions. 

Commentary: The approximate period formula indicated in Method 2 is different 

from that contained in either FEMA-273 or FEMA-302. This formula has been 

adapted from a recent study of the statistical distribution of measured periods in 

buildings obtained from accelerometer recordings of excitation occurring in past 

earthquakes (Goel and Chopra, 1997). This formula is intended to provide 

approximately an 84% confidence level (mean + one s) that the actual period will 

exceed the calculated value. The formula has intentionally been selected to 

under-estimate the actual period of the building as this will result in a 

conservatively large estimate of the calculated pseudo lateral force applied to the 

structure as a loading (see Section 4.4.3.3.1). The large pseudo lateral force will 

result in conservatively large estimates of interstory drift. 

Use of the more accurate Method 1 procedure will typically result in lower 

estimates of interstory drift, and therefore increased confidence in the ability of a 

building to meet performance goals. 

4.4.3.3 Determination of Actions and Deformations 

4.4.3.3.1 Pseudo Lateral Load 

A pseudo lateral load V, given by Equation 4-3, shall be independently calculated for each of 

the two orthogonal directions of building response, and applied to a mathematical model of the 

structure. 

V = C1C2 C3SaW (4-3) 

where: 

C1 = a modification factor to relate expected maximum inelastic displacements to 

displacements calculated for linear elastic response. C1 may be calculated using 

the procedure indicated in Section 3.3.3.3 in FEMA-273 with the elastic base 

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shear capacity substituted for Vy. Alternatively, C1 may be taken as having a value 

of 1.0 when the fundamental period T of the building response is greater than Ts 

and shall be taken as having a value of 1.5 when the fundamental period of the 

structure is equal to or less than T0. Linear interpolation shall be used to calculate 

C1 for intermediate values of T. 

T0 = period at which the acceleration response spectrum for the site reaches its 

peak value, as indicated in FEMA-302. It may be taken as 0.2Ts. 

TS = the characteristic period of the response spectrum, defined as the period 

associated with the transition from the constant spectral response 

acceleration segment of the spectrum to the constant spectral response 

velocity segment of the spectrum, as defined in FEMA-302. 

C2 = a modification factor to represent the effect of hysteretic pinching on the 

maximum displacement response. For steel moment frames the value of C2 shall 

be taken as 1.0. 

C3 = a modification factor to represent increased dynamic displacements due to P-D 

effects and stiffness degradation. C3 may be taken from Table 4-4 or shall be 

calculated from the equation: 

C3 = 1+ 

5(qi - 0.1) ‡ 1.0 (4-4) T 

where: 

qi = the coefficient calculated in accordance with Section 2.1.1.2 of FEMA273. The maximum value qi for all stories in the building shall be used. 

Sa = Response spectrum acceleration, at the fundamental period and damping ratio of 

the building in the direction under consideration, for the hazard level 

corresponding to the performance objective being evaluated (i.e., probability of 

exceedance). The value of Sa may be calculated using the procedure outlined in 

Section 2.6.1.5 of FEMA-273. 

W = Total dead load and anticipated live load as indicated below: 

• In storage and warehouse occupancies, a minimum of 25% of the floor live 

load 

• The actual partition weight or minimum weight of 10 psf of floor area, 

whichever is greater 

• The applicable snow load – see FEMA-302 

• The total weight of permanent equipment and furnishings 

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Table 4-4 Modification Factor C3 for the Linear Static Procedure 

Performance Level C3 

Immediate Occupancy 1.0 

Collapse Prevention 

Connections meeting the criteria for Special 

Moment Frame structures in accordance with 

Chapter 3 

1.2 

Connections meeting the criteria for Ordinary 

Moment Frame structures in accordance with 

Chapter 3 

1.4 

Commentary: The pseudo lateral force, when distributed over the height of the 

linear-elastic model of the structure, is intended to produce calculated lateral 

displacements approximately equal to those that are expected in the real structure 

during the design event. If it is expected that the actual structure will yield during 

the design event, the force given by Equation 4-3 may be significantly larger than 

the actual strength of the structure to resist this force. The acceptance evaluation 

procedures in Section 4.6 are developed to take this into account. 

The values of C3 in Table 4-4 are conservative for most structures, and will 

generally result in calculation of an unduly low level of confidence. Use of 

Equation 4-4 to calculate C3 is one way to improve calculated confidence without 

extensive additional effort, and is recommended. 

4.4.3.3.2 Vertical Distribution of Seismic Forces 

The lateral load Fx applied at any floor level x shall be determined as given in Section 

3.3.1.3B of FEMA-273. 

4.4.3.3.3 Horizontal Distribution of Seismic Forces 

The seismic forces at each floor level of the building shall be distributed according to the 

distribution of mass at that floor level. 

4.4.3.3.4 Diaphragms 

Floor and roof diaphragms shall be evaluated using the procedure outlined in Section 

3.3.1.3D in FEMA-273. The lateral seismic load on each flexible diaphragm shall be distributed 

along the span of that diaphragm, considering its displaced shape. 

4.4.3.3.5 Determination of Interstory Drift 

Interstory drifts shall be calculated using lateral loads in accordance with Section 4.4.3.3.1 

and stiffness obtained from Section 4.5. Factored interstory drift demands gagdi at each story i 

shall be determined by applying the appropriate analysis uncertainty factor ga and demand 

variability factor g obtained from Section 4.6. 

4-18
Œ œ 

�

� � � 

�

� � � 

Recommended Seismic Design

Moment-Frame Buildings Chapter 4: Performance Evaluation

Criteria for New Steel FEMA-350 

4.4.3.3.6 Determination of Column Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

calculated column forces by the applicable analysis uncertainty factor ga and demand variability 

factor g obtained in Section 3.6.3. Column forces shall be calculated either as: 

1. the axial demands from the unreduced linear analysis, or 

2. the axial demands computed from the equation: 

Ø � n M � � n M � ø 

P' c = –Œ2�

�� pe �� - 2�

�� pe �� œ (4-5) 

o Ł i=x L łL Ł i=x L łR ß 

where: 

n M pe �

Ł i= x L �

�

ł L 

= the summation of the expected plastic moment strength (ZFye) divided by 

the span length L, of all beams framing into the left hand side of the 

column, above the level under consideration, and 

n M pe �

Ł i= x L �

�

ł R 

= the summation of the expected plastic moment strength (ZFye) divided by 

the span length L, of all beams framing into the right hand side of the 

column, above the level under consideration. 

When a column is part of framing that resists lateral forces under multiple directions of 

loading, the Seismic Demand shall be taken as the most severe condition resulting from 

application of 100% of the Seismic Demand computed for any one direction of response with 

30% of the Seismic Demand computed for the orthogonal direction of response. 

4.4.4 Linear Dynamic Procedure 

4.4.4.1 Basis of the Procedure 

Linear dynamic procedure (LDP) analysis of steel moment frames shall be conducted in 

accordance with the recommendations in Section 3.3.2 of FEMA-273 except as specifically noted 

herein. Coefficients C1, C2, and C3 should be taken as indicated in Section 4.4.3.3 of this 

document. 

Estimates of interstory drift and column axial demands shall be evaluated using the 

applicable procedures of Section 4.6. Calculated displacements and column axial demands are 

factored by the applicable analysis uncertainty factor ga and demand variability factor g obtained 

from Section 4.6 and compared with factored capacity values for the appropriate performance 

level. Calculated internal forces typically will exceed those that the building can sustain because 

of anticipated inelastic response of components and elements, but are generally not used to 

evaluate performance. 

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Commentary: The linear dynamic procedure (LDP) is similar in approach to the 

linear static procedure (LSP), described in Section 4.4.3. However, because it 

directly accounts for the stiffness and mass distribution of the structure in 

calculating the dynamic response characteristics, its use introduces somewhat 

less uncertainty than does the LSP. Coefficients C1, C2, and C3, which account in 

an approximate manner for the differences between elastic predictions of 

response and inelastic behavior, are the same as for the linear static method. In 

the LDP, inertial seismic forces, their distribution over the height of the building, 

and the corresponding internal forces and system displacements are determined 

using a linear-elastic, response spectrum analysis. 

The basis, modeling approaches and acceptance criteria of the LDP are 

similar to those for the LSP. The main exception is that the response calculations 

are carried out using modal response spectrum analysis (RSA). Modal spectral 

analysis is carried out using unreduced, linear-elastic response spectra scaled to 

the hazard level (probability of exceedance) inherent in the desired performance 

objective. As with the LSP, it is expected that the LDP will produce estimates of 

displacements and interstory drifts that are approximately correct, but will 

produce estimates of internal forces that exceed those that would be obtained in a 

yielding building. 

4.4.4.2 Analysis 

4.4.4.2.1 General 

The LDP shall conform to the criteria in Section 3.3.2.2 of FEMA-273. The analysis shall be 

based on appropriate characterization of the ground motion. The requirement that all significant 

modes be included in the response analysis may be satisfied by including sufficient modes to 

capture at least 90% of the participating mass of the building in each of the building’s principal 

horizontal directions. Modal damping ratios should reflect the damping inherent in the building 

at deformation levels less than the yield deformation. Except for buildings incorporating passive 

or active energy dissipation devices, or base isolation technology, effective damping shall be 

taken as 5% of critical. 

The interstory drift, and other response parameters calculated for each mode, and required for 

evaluation in accordance with Section 4.4.4.3, should be combined by recognized methods to 

estimate total response. Modal combination by either the SRSS (square root of the sum of 

squares) rule or the CQC (complete quadratic combination) rule is acceptable. 

Multidirectional excitation effects may be accounted for by combining 100% of the response 

due to loading in direction A with 30% of the response due to loading in the direction B, and by 

combining 30% of the response in direction A with 100% of the response in direction B, where A 

and B are orthogonal directions of response for the building. 

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4.4.4.2.2 Ground Motion Characterization 

The horizontal ground motion should be characterized by one of the following methods: 

1. An elastic response spectrum, developed in accordance with the recommendations of Section 

2.6.1.5 in FEMA-273 for the hazard level contained in the desired performance objective. 

2. A site-specific response spectrum developed in accordance with the recommendations of 

Section 2.6.2.1 of FEMA-273 for the appropriate hazard level contained in the desired 

performance objective. 

4.4.4.3 Determination of Actions and Deformations 

4.4.4.3.1 Factored Interstory Drift Demand 

Factored interstory drift demand shall be obtained by multiplying the interstory drift results of 

the response spectrum analysis by the product of the modification factors, C1, C2, and C3, defined 

in Section 4.4.3 and by the applicable analysis uncertainty factor ga and demand variability factor 

g obtained from Section 4.6. 

4.4.4.3.2 Determination of Column Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

calculated column forces, as given in Section 4.4.3.3.6, by the applicable analysis uncertainty 

factor ga and demand variability factor g obtained from Section 4.6.3. 

4.4.5 Nonlinear Static Procedure 

4.4.5.1 Basis of the Procedure 

Under the Nonlinear Static Procedure (NSP), a model directly incorporating the inelastic 

material and nonlinear geometric response is displaced to a target displacement, and resulting 

internal deformations and forces are determined. The nonlinear load-deformation characteristics 

of individual components and elements of the building are modeled directly. The mathematical 

model of the building is subjected to a pattern of monotonically increasing lateral forces or 

displacements until either a target displacement is exceeded or a mathematical instability occurs. 

The target displacement is intended to approximate the total maximum displacement likely to be 

experienced by the actual structure, at the hazard level corresponding to the selected performance 

objective. The target displacement should be calculated in accordance with the procedure 

presented in Section 3.3.3.3 of FEMA-273 with the modifications indicated below. Because the 

mathematical model accounts directly for effects of material and geometric nonlinear response, 

the calculated internal forces will be reasonable approximations of those expected during the 

design earthquake, presuming that an appropriate pattern of loading has been applied. 

Interstory drifts and column axial demands obtained from the NSP are evaluated using the 

applicable procedures of Section 4.6. Calculated interstory drifts, column forces, and column 

splice forces are factored, and compared directly with factored acceptance values for the 

applicable performance level. 

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Commentary: The nonlinear static analysis approach inherently assumes 

behavior is dominated by the first mode response of the structure. For this 

reason, these Recommended Criteria state that this approach be used only for 

structures with relatively short periods. What constitutes a building with a “short 

period” is dependent on the spectral characteristics of ground shaking 

anticipated at the site. The small magnitude events that dominate the hazard at 

many central and eastern U.S. sites tend to have most of their energy at short 

periods, particularly on firm soil and rock sites. For sites subject to such 

shaking, nonlinear static analyses may be valid only for short-period, rigid 

structures. The limitations on use of the Nonlinear Static Procedure (NSP), based 

on period, contained in Table 4-3, are based on recent work that indicates that 

higher mode response does not become significant in structures responding to 

ground shaking having typical response spectra unless the fundamental period of 

the structure is more than about 3.5 times the period at which the spectrum 

transitions from a range of constant response acceleration to constant response 

velocity. 

A second potential limitation of this procedure is that in practice, twodimensional models are often used to simulate three-dimensional response. 

Estimates of load distribution between the lateral-load-resisting elements in the 

building are required, and the accuracy of the analysis depends upon the 

accuracy of distribution. Three-dimensional linearly elastic models may be used 

to estimate this distribution; however, these models are unable to account for 

load redistribution occurring because of inelastic behavior. When many plastic 

hinges form nearly simultaneously, creating local frame mechanisms, the effects 

of torsional contributions may not be accurately represented. If a structure has 

significant torsional irregularity, three-dimensional models should be used. 

The NSP is also limited with regard to evaluation of simultaneous response to 

ground shaking in different directions. Little research has been performed on 

appropriate methods of accounting for multi-directional response using this 

technique. Therefore, these Recommended Criteria have adapted standard 

approaches used in linear analysis for this purpose. 

4.4.5.2 Analysis Considerations 

4.4.5.2.1 General 

In the context of these Recommended Criteria, the NSP involves the application of 

incrementally adjusted, monotonically increasing, lateral forces, or displacements, to a 

mathematical nonlinear model of a building, until the displacement of a control node in the 

mathematical model exceeds a target displacement. For buildings that are not symmetric about a 

plane perpendicular to the applied lateral loads, the lateral loads must be applied in both the 

positive and negative directions, and the maximum forces and deformations obtained from both 

directions used for design. 

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The relation between base shear force and lateral displacement of the control node should be 

established for control node displacements ranging between zero and 150% of the target 

displacement dt given by Equation 3-11 of FEMA-273. Performance evaluation should be based 

on those column forces and interstory drifts corresponding to minimum horizontal displacement 

of the control node equal to the target displacement dt corresponding to the hazard level 

appropriate to the performance objective being evaluated. 

Gravity loads shall be applied to appropriate elements and components of the mathematical 

model during the Nonlinear Static Procedure (NSP). The loads and load combinations shall be as 

follows: 

1. 100% of computed dead loads and permanent live loads shall be applied to the model. 

2. 25% of transient floor live loads shall be applied to the model, except in warehouse and 

storage occupancies, where the percentage of live load used in the analysis shall be based on 

a realistic assessment of the average long-term loading. 

The analysis model should be discretized in sufficient detail to represent adequately the loaddeformation response of each component along its length. Particular attention should be paid to 

identifying locations of inelastic action along the length of a component, as well as at its ends. 

Commentary: As with any nonlinear model, the ability of the analyst to detect the 

presence of inelastic behavior requires the use of a nonlinear finite element at the 

assumed location of yielding. The model will fail to detect inelastic behavior when 

appropriately distributed finite elements are not used. However, as an alternative 

to the use of nonlinear elements, it is possible to use linear elements and 

reconfigure the model, for example, by adjusting member restraints, as 

nonlinearity is predicted to occur. For example, when a member is predicted to 

develop a plastic hinge, a linear model can be revised to place a hinge at this 

location. When this approach is used, the internal forces and stresses that caused 

the hinging must be reapplied as a nonvarying static load. 

The recommendation to continue the pushover analysis to displacements that 

are 150% of the target displacement is to allow an understanding of the probable 

behavior of the building under somewhat larger loading than anticipated. If the 

pushover analysis should become unstable prior to reaching 150% of the target 

displacement, this does not indicate that a design is unacceptable, but does 

provide an indication of how much reserve remains in the structure at the design 

ground motion. 

4.4.5.2.2 Control Node 

The NSP requires definition of a control node in the building. These Recommended Criteria 

consider the control node to be the center of mass at the roof of the building; the top of a 

penthouse should not be considered as the roof unless it is of such substantial area and 

construction as to materially affect the response. The displacement of the control node is 

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compared with the target displacement – a displacement that characterizes the effects of 

earthquake shaking at the desired hazard level. 

4.4.5.2.3 Lateral Load Patterns 

Lateral loads should be applied to the building in profiles given in Section 3.3.3.2C of 

FEMA-273. 

4.4.5.2.4 Period Determination 

The effective fundamental period Te in the direction under consideration shall be calculated 

using the force-displacement relationship of the Nonlinear Static Procedure (NSP) as described 

in Section 3.3.3.2D of FEMA-273. 

4.4.5.2.5 Analysis of Three-Dimensional Models 

Static lateral forces shall be imposed on the three-dimensional mathematical model 

corresponding to the mass distribution at each floor level. 

Independent analysis along each principal axis of the three-dimensional mathematical model 

is permitted unless multidirectional evaluation is required by Section 3.2.7 in FEMA-273. Refer 

also to Section 4.4.5.3.4 of these Recommended Criteria. 

4.4.5.2.6 Analysis of Two-Dimensional Models 

Mathematical models describing the framing along each axis of the building should be 

developed for two-dimensional analysis. The effects of horizontal torsion should be considered 

as required by Section 3.2.2.2 of FEMA-273. 

4.4.5.3 Determination of Actions and Deformations 

4.4.5.3.1 Target Displacement 

The target displacement dt for buildings with rigid diaphragms at each floor level shall be 

estimated using the procedures of Section 3.3.3.3A of FEMA-273. Actions and deformations 

corresponding to the control node displacement equal to the target displacement shall be used for 

evaluation of performance evaluation in accordance with Section 4.6. 

4.4.5.3.2 Diaphragms 

The lateral seismic load on each flexible diaphragm shall be distributed along the span of that 

diaphragm, considering its displaced shape. 

4.4.5.3.3 Factored Interstory Drift Demand 

Factored interstory drift demand shall be obtained by multiplying the maximum interstory 

drift calculated at the target displacement by the applicable analysis uncertainty factor ga and 

demand variability factor g obtained from Section 4.6. 

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4.4.5.3.4 Multidirectional Effects 

Multidirectional excitation effects may be accounted for by combining 100% of the response 

due to loading in direction A with 30% of the response due to loading in the direction B; and by 

combining 30% of the response in direction A with 100% of the response in direction B, where A 

and B are orthogonal directions of response for the building. 

An acceptable alternative to this approach is to perform the Nonlinear Static Procedure (NSP) 

analysis simultaneously in two orthogonal directions by application of 100% of the loading in 

direction A simultaneously with 30% of the loading in direction B. Loading shall be applied 

until 100% of the target displacement in direction A is achieved. This procedure shall be 

repeated with 100% of the loading applied in direction A and 30% in direction B. 

4.4.5.3.5 Factored Column and Column Splice Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

calculated column forces at the target displacement by the applicable analysis uncertainty factor 

ga and demand variability factor g from Section 4.6. 

4.4.6 Nonlinear Dynamic Procedure 

4.4.6.1 Basis of the Procedure 

Under the Nonlinear Dynamic Procedure (NDP), inertial seismic forces, their distribution 

over the height of the building, and the corresponding internal forces and system displacements 

are determined using an inelastic response-history dynamic analysis. 

The basis, the modeling approaches, and the acceptance criteria for the NDP are similar to 

those for the NSP. The main exception is that the response calculations are carried out using 

response-history analysis. With the NDP, the design displacements are not established using a 

target displacement, but instead are determined directly through dynamic analysis using suites of 

ground motion records. Calculated response can be highly sensitive to characteristics of 

individual ground motions; therefore, it is necessary to carry out the analysis with more than one 

ground motion record. Because the numerical model accounts directly for effects of material and 

geometric inelastic response, the calculated internal forces will be reasonable approximations of 

those expected during the design earthquake. 

Results of the NDP are to be checked using the applicable acceptance criteria of Section 4.6. 

Calculated displacements and internal forces are factored, and compared directly with factored 

acceptance values for the applicable performance level. 

4.4.6.2 Analysis Assumptions 

4.4.6.2.1 General 

The NDP shall conform to the criteria given in Section 3.3.4.2A of FEMA-273. 

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4.4.6.2.2 Ground Motion Characterization 

The earthquake shaking should be characterized by suites of ground motion acceleration 

histories, prepared in accordance with the recommendations of Section 2.6.2 of FEMA-273 and 

corresponding to the hazard level appropriate to the desired performance objective. A minimum 

of three pairs of ground motion records shall be used. Each pair shall consist of two orthogonal 

components of the ground motion. 

Consideration of multidirectional excitation effects required by Section 3.2.7 of FEMA-273 

may be satisfied by analysis of a three-dimensional mathematical model using simultaneously 

imposed pairs of earthquake ground motion records along the horizontal axes of the building. 

The effects of torsion should be considered according to Section 3.2.2.2 of FEMA-273. 

4.4.6.3 Determination of Actions and Deformations 

4.4.6.3.1 Response Quantities 

Response quantities shall be computed as follows: 

1. If less than seven pairs of ground motion records are used to perform the analyses, each 

response quantity (for example, interstory drift demand, or column axial demand) shall be 

taken as the maximum value obtained from any of the analyses. 

2. If seven or more pairs of ground motion records are used to perform the analyses, the median 

value of each of the response quantities computed from the suite of analyses may be used as 

the demand. The median value shall be that value exceeded by 50% of the analyses in the 

suite. 

4.4.6.3.2 Factored Interstory Drift Demand 

Factored interstory drift demand shall be obtained by multiplying the maximum of the 

interstory drifts calculated in accordance with Section 4.4.6.3.1 by the applicable analysis 

uncertainty factor ga and demand variability factor g obtained from Section 4.6. 

4.4.6.3.3 Factored Column and Column Splice Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

column forces calculated in accordance with Section 4.4.6.3.1 by the applicable analysis 

uncertainty factor ga and demand variability factor g from Section 4.6. 

4.5 Mathematical Modeling 

4.5.1 Basic Assumptions 

In general, a steel moment-frame building should be modeled and analyzed as a threedimensional assembly of elements and components. Although two-dimensional models may 

provide adequate response information for regular, symmetric structures and structures with 

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flexible diaphragms, three-dimensional mathematical models should be used for analysis and 

design of buildings with plan irregularity as defined in FEMA-302. Two-dimensional modeling, 

analysis, and design of buildings with stiff or rigid diaphragms is acceptable, if torsional effects 

are either sufficiently small to be ignored, or are captured indirectly. 

Vertical lines of framing in buildings with flexible diaphragms may be individually modeled, 

analyzed and designed as two-dimensional assemblies of components and elements, or a threedimensional model may be used with the diaphragms modeled as flexible elements. 

Explicit modeling of connection force-deformation behaviors for fully restrained connections 

is not required for linear analysis procedures. The stiffness of partially restrained connections 

should be modeled in linear procedures in accordance with the recommendations of Section 

4.5.2. In nonlinear procedures explicit modeling of connection stiffness is recommended for 

those cases when the connection is weaker than the connected components, or when it is 

appropriate to model strength degradation in the connection as a function of imposed 

deformation demand. Refer to Section 4.5.2. 

Commentary: A finite element model will only collect information at places in the 

structure where a modeling element is inserted. When nonlinear deformations are 

expected in a structure, the analyst must anticipate the location of these 

deformations (such as plastic hinges) and insert nonlinear finite elements at these 

locations if the inelastic behavior is to be captured by the model. 

4.5.2 Frame Configuration 

The analytical model should accurately account for the stiffness of frame elements and 

connections. Element and component stiffness properties, strength estimates and locations of 

plastic hinge formation for both linear and nonlinear procedures can be determined from 

information given in Chapter 3 for prequalified connections. 

4.5.2.1 Modeling 

Only the beams and columns forming the lateral-force-resisting system need be modeled. 

However, it shall be permissible to model nonparticipating elements of the structure if realistic 

assumptions are made with regard to their stiffness, strength and deformation capacity. 

Commentary: Analyses of buildings for the purposes of demonstrating 

compliance with the strength and drift criteria of FEMA-302 must neglect the 

participation of gravity-load-carrying beams and columns that are not intended to 

be part of the lateral-force-resisting system. Studies conducted in support of the 

development of these Recommended Criteria indicate that these connections are 

capable of contributing non-negligible stiffness through large interstory drift 

demands. Analyses made with models that neglect the participation of these 

elements will tend to over-estimate demands on the lateral-force-resisting 

elements and interstory drift demand on the structure. The demand factors 

provided in Section 4.6 have been calibrated to account for this over-estimation. 

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While it is permissible to conduct performance evaluations using models that 

neglect nonparticipating framing, models that include the stiffness of these 

elements can be used to provide improved levels of confidence with regard to the 

building’s ability to meet desired performance objectives. This is an example of 

the process by which confidence can be improved – by performing more intense 

study to reduce the inherent uncertainty. 

4.5.2.2 Connection Modeling 

4.5.2.2.1 Fully Restrained Moment-Resisting Connections 

Elastic analysis models of structures with fully restrained connections should be based on the 

assumption that the connection provides a fully rigid interconnection between the beam and 

column, located at the centerline of the column. Alternatively, realistic assumptions with regard 

to panel zone flexibility may be made, as indicated in Section 4.5.2.3. 

Nonlinear analysis models of structures with fully restrained connections should be based on 

the assumption that the connection provides a fully rigid interconnection between the beam and 

column, located at the centerline of the column, until either the connection panel zone, beam or 

column yields, or a total interstory drift angle qSD (obtained from Table 4-12) occurs. The 

expected yield strength of the material, as indicated in Section 2.6.2 should be used to calculate 

the yield capacity of beams, columns, and panel zones. If yielding occurs at total interstory drift 

angles less than qSD, the yielding element should be assumed to exhibit plastic behavior. At 

interstory drifts greater than qSD the connection should be assumed to be capable of transmitting 

20% of the expected plastic moment capacity of the girder until a total interstory drift angle qU, 

(also obtained from Table 4-12) occurs. At interstory drift angles greater than qU, the connection 

should be presumed to have negligible strength. 

4.5.2.2.2 Partially Restrained Moment-Resisting Connections 

Models of frames incorporating partially restrained connections should explicitly account for 

the stiffness of the connection. For linear models, connection stiffness may be modeled by 

incorporating a rotational spring element between the beam and column. Alternatively, a 

modified beam with partially restrained connections may be modeled as rigidly attached to 

columns, and using an effective modulus of rigidity, EIeq, for the beam that accounts for the 

reduced stiffness introduced by the connection. For beams with similar partially restrained 

connections on each end, the effective modulus of rigidity may be calculated as: 

1 EIeq = 6h 1 (4-6) 

+ 

l

2 K EI b q b 

where: 

E = the modulus of elasticity, kip/ square inch 

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h = the average story height of the columns above and below the beam, inches

Ib = the moment of inertia of the beam, (inches)4

lb = the beam span center to center of columns, inches

Kq = the stiffness of the connection, kip-in/radian

Refer to Section 3.7.1 for recommended connection stiffness for Double Split Tee partially 

restrained connections. Stiffness for other partially restrained connections should be based on 

laboratory data or rational analysis. 

For nonlinear analysis, the connection should be explicitly modeled as an elastic-perfectlyplastic nonlinear spring with an elastic stiffness calculated as indicated above, and a plastic 

strength equal to the expected strength of the yield mode for the connection. Section 3.7.1 

provides recommendations for determining the expected strength of the yield mode for Double 

Split Tee partially restrained connections. Expected strength of other types of partially restrained 

connections should be based on laboratory data or rational analysis. Partially restrained 

connections should be assumed to have negligible strength at interstory drift angle demands that 

exceed qu, as indicated in Section 4.6. 

4.5.2.2.3 Simple Shear Tab Connections 

When included in linear analytical models the stiffness of simple shear tab connections 

should be explicitly modeled as a rotational spring that connects the beam to the column. The 

spring stiffness, Kq should be taken as: 

Kq = 28000(dbg - 5.6) (4-7) 

where dbg is the bolt group depth in inches and Kq is in units of k-inches per radian. In lieu of 

explicit modeling of the connection, beams that frame into columns with simple shear tab 

connections may be modeled with an equivalent rigidity, EIeq calculated in accordance with 

Equation 4-6, of Section 4.5.2.2.2. 

When simple shear tab connections are included in nonlinear analysis models, they should be 

explicitly modeled as an elastic-perfectly-plastic rotational spring. The elastic stiffness of the 

spring should be taken as given by Equation 4-7. The plastic strength of the spring should be 

determined as the expected plastic moment capacity of the bolt group, calculated as the sum of 

the expected yield strength of the bolts and their distance from the neutral axis of the bolt group. 

The expected yield strength shall be taken as125% of the capacity of the bolt group determined in 

accordance with AISC LRFD using a resistance factor f of unity. Simple shear tab connections 

should be assumed to have negligible strength at interstory drift angle demands that exceed qu, as 

indicated in Section 4.6. 

4.5.2.3 Panel Zone Stiffness 

It shall be permissible for the model to assume centerline-to-centerline dimensions for the 

purpose of calculating stiffness of beams and columns. Alternatively, more realistic assumptions 

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that account for the rigidity of panel zones may be used. Regardless, calculation of moments and 

shears should be performed at the face of the column. 

Commentary: Models that use centerline-to-centerline dimensions for calculation 

of beam and column stiffness will tend to overestimate the interstory drift demand 

on the structure. The demand factors provided in Section 4.6 have been calibrated 

to account for this overestimation. While it is permissible to conduct performance 

evaluations using models that neglect the stiffening effect of the panel zone on 

beam and column stiffness, models that include more realistic estimation of this 

stiffness can be used to provide improved levels of confidence with regard to the 

building’s ability to meet desired performance objectives. 

A number of models are available to represent panel zones of momentresisting connections. These range from simple models that treat the panel zone 

as a series of rigid links extending outward from the center of the beam-column 

connection and along the axes of the beams and columns to scissors-type models 

that explicitly account for the shear stiffness of the panel zone, to complex multielement models that account both for shear stiffness of the panel zone and the 

effects of geometric distortion of the zone. Analyses of buildings using these 

various models, reported in FEMA-355C indicate that the particular model used 

has relatively little impact on the predicted interstory drift demand. However, for 

nonlinear analysis models, the element selected to represent the panel zone can 

have significant impact on where plasticity in the structure is predicted to occur, 

e.g., in the panel zone itself, within the beam, or a combination of these regions. 

4.5.3 Horizontal Torsion 

The effects of actual horizontal torsion must be considered. In FEMA-302, the total torsional 

moment at a given floor level includes the following two torsional moments: 

a. the actual torsion, that is, the moment resulting from the eccentricity between the centers of 

mass at all floors above and including the given floor, and the center of rigidity of the vertical 

seismic elements in the story below the given floor, and 

b. the accidental torsion, that is, an accidental torsional moment produced by horizontal offset 

in the centers of mass, at all floors above and including the given floor, equal to a minimum 

of 5% of the horizontal dimension at the given floor level measured perpendicular to the 

direction of the applied load. 

For the purposes of performance evaluation, under these Recommended Criteria, accidental 

torsion should not be considered. In buildings with diaphragms that are not flexible, the effect of 

actual torsion should be considered if the maximum lateral displacement dmax from this effect at 

any point on any floor diaphragm exceeds the average displacement davg by more than 10%. 

Commentary: Accidental torsion is an artificial device used by the building codes 

to account for actual torsion that can occur, but is not apparent in an evaluation 

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of the center of rigidity and center of mass in an elastic stiffness evaluation. Such 

torsion can develop during nonlinear response of the structure if yielding 

develops in an unsymmetrical manner in the structure. For example if the frames 

on the east and west sides of a structure have similar elastic stiffness the structure 

may not have significant torsion during elastic response. However, if the frame 

on the east side of the structure yields significantly sooner than the framing on the 

west side, then inelastic torsion will develop. Rather than requiring that an 

accidental torsion be applied in the analysis, as do the building codes, these 

Recommended Criteria directly account for the uncertainty related to these 

torsional effects in the calculation of demand and resistance factors. Accidental 

torsion should be applied in analyses applied to the design of frames, as required 

by FEMA-302. 

4.5.4 Foundation Modeling 

In general, foundations should be modeled as unyielding. Assumptions with regard to the 

extent of fixity against rotation provided at the base of columns should realistically account for 

the relative rigidities of the frame and foundation system, including soil compliance effects, and 

the detailing of the column base connections. For purposes of determining building period and 

dynamic properties, soil-structure interaction may be modeled, as permitted by the building code. 

Commentary: Most steel moment frames can be adequately modeled by assuming 

that the foundation provides rigid support for vertical loads. However, the 

flexibility of foundation systems (and the attachment of columns to those systems) 

can significantly alter the flexural stiffness at the base of the frame. Where 

relevant, these factors should be considered in developing the analytical model. 

4.5.5 Diaphragms 

Floor and roof diaphragms transfer earthquake-induced inertial forces to vertical elements of 

the seismic framing system. Connections between floor and roof diaphragms and vertical 

seismic framing elements must have sufficient strength to transfer the maximum calculated 

diaphragm shear forces to the vertical framing elements. Requirements for evaluation of 

diaphragm components are given in Section 3.3 of FEMA-273. 

Development of the mathematical model should reflect the stiffness of the diaphragm. As a 

general rule, most floor slabs with concrete fill over metal deck may be considered to be rigid 

diaphragms and floors or roofs with plywood diaphragms should be considered flexible. The 

flexibility of unfilled metal deck, and concrete slab diaphragms with large openings should be 

considered in the analytical model. 

Mathematical models of buildings with diaphragms that are not rigid should be developed 

considering the effects of diaphragm flexibility. For buildings with flexible diaphragms at each 

floor level, the vertical lines of seismic framing may be designed independently, with seismic 

masses assigned on the basis of tributary area. 

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4.5.6 P-D Effects 

P-D effects, caused by gravity loads acting on the displaced configuration of the structure, 

may be critical in the seismic performance of steel moment-frame buildings, which are usually 

flexible and may be subjected to large lateral displacements. 

The structure should be evaluated for P-D effects in accordance with the requirements of 

Section 2.8.6 of these Recommended Criteria. Where the quantity Yi in any story calculated in 

accordance with Section 2.8.6 exceeds 0.1, the increased deflections resulting from P-D effects 

must be determined. Where the quantity Yi in any story exceeds 0.3, the interstory drift capacity 

of the structure must be determined in accordance with Appendix A of these Recommended 

Criteria. 

Commentary: The values of interstory drift capacity for the Collapse Prevention 

performance level, provided in Section 4.6, and the corresponding resistance 

factors, were computed considering P-D effects (FEMA-355F). For a given 

structure, it is believed that if the value of Y is less than 0.3 the effects of P-D 

have been adequately considered by these general studies. For values of Y 

greater than this limit, the statistics on frame interstory drift capacities in Section 

4.6 are inappropriate. For such frames explicit determination of interstory drift 

capacities by considering P-D effects, and by using the detailed performance 

evaluation procedures outlined in Appendix A is required. 

4.5.7 Multidirectional Excitation Effects 

Buildings should be evaluated for response due to seismic forces in any horizontal direction. 

For regular buildings, seismic displacements and forces may be assumed to act nonconcurrently 

in the direction of each principal axis of a building. For buildings with plan irregularity and 

buildings in which one or more components form part of two or more intersecting elements, 

multidirectional excitation effects should be considered, as indicated in Section 4.4 for the 

various analytical procedures. 

4.5.8 Vertical Ground Motion 

The effects of vertical excitation on horizontal cantilevers should be considered by static or 

dynamic response methods. Vertical earthquake shaking may be characterized by a spectrum 

with ordinates equal to 2/3 of those of the horizontal spectrum unless alternative vertical 

response spectra are developed using site-specific analysis. Vertical earthquake effects on other 

beam elements and column elements need not be considered. 

Commentary: There is no evidence that response to the vertical component of 

ground shaking has had any significant effect on the performance of steel 

moment-frame buildings. Consequently, the effect of this response is not 

recommended for consideration in performance evaluation, except as required by 

the building code for the case of horizontal cantilever elements. 

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Moment-Frame Buildings Chapter 4: Performance Evaluation

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Traditionally, vertical response spectra, when considered, have been taken as 

2/3 of the horizontal spectra developed for the site. While this is a reasonable 

approximation for most sites, vertical response spectra at near-field sites, located 

within a few kilometers of the zone of fault rupture, can have substantially 

stronger vertical response spectra than indicated by this approximation. 

Development of site-specific response spectra is recommended when vertical 

response must be considered for buildings on such sites. 

4.6 Acceptance Criteria 

Acceptability of building performance should be evaluated by determining a level of 

confidence in the building’s ability to meet the desired performance objective(s). The parameters 

in Table 4-5 must be independently evaluated, using the procedures of Section 4.6.1 and the 

parameters and acceptance criteria of Sections 4.6.2, 4.6.3 and 4.6.4, for each performance 

objective evaluated. The controlling parameter is that which results in the calculation of the 

lowest confidence for building performance. 

Table 4-5 Performance Parameters Requiring Evaluation of Confidence 

Parameter Discussion 

Interstory Drift The maximum interstory drift computed for any story of the structure shall be 

evaluated for global and local behaviors (for Collapse Prevention and 

Immediate Occupancy). Refer to Section 4.6.2. 

Column Axial Load The adequacy of each column to withstand its calculated maximum 

compressive demand shall be evaluated both for Collapse Prevention and 

Immediate Occupancy. Refer to Section 4.6.3. 

Column Splice Tension The adequacy of column splices to withstand their calculated maximum 

tensile demands shall be evaluated both for Collapse Prevention and 

Immediate Occupancy. Refer to Section 4.6.4. 

4.6.1 Factored-Demand-to-Capacity Ratio 

Confidence level is determined through evaluation of the factored-demand-to-capacity ratio 

given by the equation: 

l = gg aD (4-8) fC 

where: 

C = capacity of the structure, as indicated in sections 4.6.2, 4.6.3, and 4.6.4, for interstory 

drift demand, column compressive demand and column splice tensile demand, 

respectively, 

D = calculated demand for the structure, obtained from structural analysis, 

g = a demand variability factor that accounts for the variability inherent in the prediction 

of demand related to assumptions made in structural modeling and prediction of the 

character of ground shaking as indicated in Sections 4.6.2, 4.6.3, and 4.6.4, for 

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interstory drift demand, column compressive demand and column tensile demand, 

respectively, 

ga = an analysis uncertainty factor that accounts for bias and uncertainty, inherent in the 

specific analytical procedure used to estimate demand as a function of ground shaking 

intensity as indicated in Section 4.6.2, 4.6.3 and 4.6.4, for interstory drift demand, 

column compressive demand and column tensile demand, respectively, 

f = a resistance factor that accounts for the uncertainty and variability, inherent in the 

prediction of structural capacity as a function of ground shaking intensity, as indicated 

in Section 4.6.2, 4.6.3 and 4.6.4 for interstory drift demand, column compressive 

demand, and column tensile demand, respectively, and 

l = a confidence index parameter from which a level of confidence can be obtained. See 

Table 4-7. 

Factored demand to capacity ratio l shall be calculated using Equation 4-8 for each of the 

performance parameters indicated in Table 4-5, which also references the appropriate Section of 

this document where the various parameters, ga, g, f required to perform this evaluation may be 

found. These referenced sections also define an uncertainty parameter bUT associated with the 

evaluation of global and local interstory drift capacity, column compressive capacity, and column 

splice tensile capacity, respectively. These uncertainties are related to the building’s 

configuration, the structural framing system (Ordinary Moment Frame or Special Moment 

Frame), the type of analytical procedure employed and the performance level being evaluated. 

Table 4-6 indicates the level of confidence associated with various values of the factored demand 

to capacity ratio l calculated using Equation 4-8, for various values of the uncertainty parameter 

bUT. Linear interpolation between the values given in Table 4-6 may be used for values of 

factored demand to capacity ratio l and uncertainty bUT intermediate to those tabulated. 

Table 4-7 provides minimum recommended levels of confidence for each of the potential 

controlling behavior modes, that is, global stability, local connection capacity, column buckling 

or column splice tensile failure, for the Immediate Occupancy and Collapse Prevention 

performance levels, respectively. 

Commentary: In order to predict structural performance, these procedures rely 

on the application of structural analysis and laboratory test data to predict the 

behavior of real structures. However, there are a number of sources of 

uncertainty inherent in the application of analysis and test data to performance 

prediction. For example, the actual strength of structural materials, the quality of 

individual welded joints, and the amount of viscous damping present is never 

precisely known, but can have impact on both the actual amount of demand 

produced on the structure and its elements and their capacity to resist these 

demands. If the actual values of these and other parameters that affect structural 

performance were known, it would be possible to accurately predict both demand 

and capacity. However, this is never the case. In these procedures, confidence is 

used as a measure of the extent that predicted behavior, based on assumed 

conditions, is likely to represent reality. 

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l b

b

b

b

b

b

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Moment-Frame Buildings Chapter 4: Performance Evaluation

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Table 4-6 Confidence Levels for Various Values of l, Given bUT 

Confidence Level 10 20 30 40 50 60 70 80 90 95 99 

bUT = 0.2 

l 1.37 1.26 1.18 1.12 1.06 1.01 0.96 0.90 0.82 0.76 0.67 

bUT = 0.3 

l 1.68 1.48 1.34 1.23 1.14 1.06 0.98 0.89 0.78 0.70 0.57 

bUT = 0.4 

l 2.12 1.79 1.57 1.40 1.27 1.15 1.03 0.90 0.76 0.66 0.51 

bUT = 0.5 

l 2.76 2.23 1.90 1.65 1.45 1.28 1.12 0.95 0.77 0.64 0.46 

bUT = 0.6 

l 3.70 2.86 2.36 1.99 1.72 1.48 1.25 1.03 0.80 0.64 0.43 

Table 4-7 Recommended Minimum Confidence Levels 

Behavior Performance Level 

Immediate Occupancy Collapse Prevention 

Global Behavior Limited by Interstory Drift 50% 90% 

Local Connection Behavior Limited by 

Interstory Drift 

50% 50% 

Column Compression Behavior 50% 90% 

Column Splice Tension Behavior 50% 50% 

The extent of confidence inherent in a performance prediction is related to the 

possible variation in the several factors that affect structural demand and 

capacity, such as stiffness, damping, connection quality, and the analytical 

procedures employed. In this project, evaluations were made of the potential 

distribution of each of these factors and the effect of variation in these factors on 

the calculated value of structural demand and capacity. Each of these sources of 

uncertainty in structural demand and capacity prediction were characterized as 

part of the supporting research for this project, by a coefficient of variation, bU. 

The coefficient, bUT is the total coefficient of variation, considering all sources of 

uncertainty. It is used, together with other factors to calculate the demand and 

resistance factors. This assumes that demand and capacity are lognormally 

distributed relative to these uncertain parameters. This allows confidence to be 

calculated as a function of the number of standard deviations that factored 

demand-to-capacity-ratio, l, lies above or below a mean value. Table 4-6 

provides a solution for this calculation, using a conservative estimate of the 

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Chapter 4: Performance Evaluation Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

hazard parameter, k=3.0, that is representative of the typical seismicity of coastal 

California. Further information on this method may be found in Appendix A. 

Appendix A also provides values of l that are more appropriate to other 

conditions of seismicity, and that may be used to provide increased estimates of 

confidence in other regions. 

4.6.2 Performance Limited by Interstory Drift Angle 

4.6.2.1 Factored Interstory Drift Angle Demand 

Factored interstory drift demand should be computed as the quantity, g gaD where the demand 

D is the largest interstory drift computed from structural analysis, ga is the coefficient obtained 

from Table 4-8 and g is the coefficient obtained from Table 4-9. 

Commentary: Several structural response parameters are used to evaluate 

structural performance. The primary parameter is interstory drift. Interstory 

drift is an excellent parameter for judging the ability of a structure to resist P-D 

instability and collapse. It is also closely related to plastic rotation demand, or 

drift angle demand, on individual beam-column connection assemblies, and is 

therefore a good predictor of the performance of beams, columns and 

connections. For tall slender structures, a significant portion of interstory drift is 

a result of axial elongation and shortening of the columns. Although modeling of 

the structure should account for this frame flexibility, that portion of interstory 

drift resulting from axial column deformation in stories below the story under 

consideration may be neglected in determining local connection performance. 

This portion of the drift must be determined manually as most computer programs 

do not calculate this quantity separately. 

Table 4-8 Interstory Drift Angle Analysis Uncertainty Factors ga 

Analysis Procedure LSP LDP NSP NDP 

System Characteristic I.O.1 C.P.2 I.O.1 C.P.2 I.O.1 C.P.2 I.O.1 C.P.2 

Special Moment Frames 

Low Rise (<4 stories) 0.94 0.70 1.03 0.83 1.13 0.89 1.02 1.03 

Mid Rise (4-12 stories) 1.15 0.97 1.14 1.25 1.45 0.99 1.02 1.06 

High Rise (> 12 stories) 1.12 1.21 1.21 1.14 1.36 0.95 1.04 1.10 

Ordinary Moment Frames 

Low Rise (<4 stories) 0.79 0.98 1.04 1.32 0.95 1.31 1.02 1.03 

Mid Rise (4-12 stories) 0.85 1.14 1.10 1.53 1.11 1.42 1.02 1.06 

High Rise (> 12 stories) 0.80 0.85 1.39 1.38 1.36 1.53 1.04 1.10 

Notes: 1. I.O. = Immediate Occupancy Performance Level 

2. C.P. = Collapse Prevention Performance Level 

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Moment-Frame Buildings Chapter 4: Performance Evaluation

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Table 4-9 Interstory Drift Angle Demand Variability Factors g 

Building 

Height 

g 

Immediate 

Occupancy 

(I.O.) 

Collapse 

Prevention (C.P.) 

Special Moment Frame 

Low Rise 

(1 - 3 stories) 

1.5 1.3 

Mid Rise 

(4 - 12 stories) 

1.4 1.2 

High Rise 

(> 12 stories) 

1.4 1.5 

Ordinary Moment Frame 

Low Rise 

(1 - 3 stories) 

1.4 1.4 

Mid Rise 

(4 - 12 stories) 

1.3 1.5 

High Rise 

(> 12 stories) 

1.6 1.8 

4.6.2.2 Factored Interstory Drift Angle Capacity 

Interstory drift capacity may be limited either by the global response of the building, or by the 

local behavior of beam-column connections. Section 4.6.2.2.1 provides values for global 

interstory drift capacity for regular, well-configured structures as well as associated uncertainties, 

bUT. Global interstory drift capacities for irregular structures must be determined using the 

detailed procedures of Appendix A. Section 4.6.2.2.2 provides values for local interstory drift 

capacity of prequalified Special Moment Frame (SMF) and Ordinary Moment Frame (OMF) 

connections. Local interstory drift capacities for connections that are not prequalified in Chapter 

3 of these Recommended Criteria must be determined in accordance with the detailed procedures 

of Appendix A. 

4.6.2.2.1 Global Interstory Drift Angle 

Factored interstory drift angle capacity fC as limited by global response of the building, shall 

be based on the product of the resistance factor f and capacity C, obtained from Table 4-10. 

Table 4-11 provides values of the uncertainty coefficient bUT to be used with global interstory 

drift evaluation. 

4.6.2.2.2 Local Interstory Drift Angle 

Factored interstory drift angle capacity fC, limited by local connection response, shall be 

based on the capacity C of the connection and the resistance factor f taken from Table 4-12 for 

prequalified connections. In Table 4-12 the capacity C for Collapse Prevention is the interstory 

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f f

b

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Chapter 4: Performance Evaluation Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

drift angle qU while capacity C for Immediate Occupancy is the interstory drift angle qIO . For 

connection types not contained in the Table 4-12, the more detailed procedures of Appendix A 

should be used to determine interstory drift angle capacity. Table 4-13 provides values of the 

uncertainty coefficient bUT to be used with local interstory drift evaluation. 

Table 4-10 Global Interstory Drift Angle Capacity C and Resistance Factors f for Regular 

SMF and OMF Buildings 

Building Height Performance Level 

Immediate Occupancy Collapse Prevention 

Interstory 

Drift Angle 

Capacity C 

Resistance 

Factor f 

Interstory 

Drift Angle 

Capacity C 

Resistance 

Factor f 

Special Moment Frames (SMF) 

Low Rise (3 stories or less) 0.02 1.0 0.10 0.90 

Mid Rise ( 4 – 12 stories) 0.02 1.0 0.10 0.85 

High Rise (> 12 stories) 0.02 1.0 0.085 0.75 

Ordinary Moment Frames (OMF) 

Low Rise (3 stories or less) 0.01 1.0 0.10 0.85 

Mid Rise ( 4 – 12 stories) 0.01 0.9 0.08 0.70 

High Rise (> 12 stories) 0.01 0.85 0.06 0.60 

Table 4-11 Uncertainty Coefficient bUT for Global Interstory Drift Evaluation 

Building Height Performance Level 

Immediate Occupancy Collapse Prevention 

Special Moment Frames (SMF) 

Low Rise (3 stories or less) 0.20 0.3 

Mid Rise ( 4 – 12 stories) 0.20 0.4 

High Rise (> 12 stories) 0.20 0.5 

Ordinary Moment Frames (OMF) 

Low Rise (3 stories or less) 0.20 0.35 

Mid Rise ( 4 – 12 stories) 0.20 0.45 

High Rise (> 12 stories) 0.20 0.55 

Notes: 1- Value of bUT should be increased by 0.05 for linear static analysis 

2- Value of bUT may be reduced by 0.05 for nonlinear dynamic analysis 

Commentary: Table 4-12 presents median drift angle capacities C indicated by 

qIO, qSD, and qU and resistance factors f for connections that are prequalified 

under Chapter 3 of these Recommended Criteria. These values were determined 

from cyclic tests of full-size connection assemblies using the testing protocols 

indicated in Chapter 3. The cyclic tests are used to determine the load4-38
q q

q

q q

Recommended Seismic Design

Moment-Frame Buildings Chapter 4: Performance Evaluation

Criteria for New Steel FEMA-350 

deformation hysteresis behavior of the system and to determine the statistics 

(medians and standard deviations) for the connection drift angle at which the 

following behaviors occur: 

1. onset of local flange buckling of beams, 

2. degradation of moment-resisting capacity of the assembly to a value below the 

nominal moment-resisting capacity, 

3. initiation of fracture of bolts, welds, or base metal that results in significant strength 

degradation of the assembly, 

4. complete failure of the connection, characterized by an inability of the connection to 

maintain integrity of the beam-to-column connection under gravity loading. 

Based on these data, limiting drift angle capacities have been obtained for the 

Immediate Occupancy and Collapse Prevention damage states, as indicated in 

Table 4-14. 

Table 4-12 Drift Angle Capacity C (qIO, qU) for Prequalified Connections as Limited By 

Local Connection Response 

Connection 

Type 

Strength 

Degradation 

Limit Drift Angle 

(radians) 

qSD 

Immediate Occupancy Collapse Prevention 

Limit Drift 

Angle 

(radians) 

qIO 

Capacity 

Reduction 

Factor 

f 

Limit Drift 

Angle 

(radians) 

qU 

Capacity 

Reduction 

Factor 

f 

WUF-B 0.031-0.0003db 0.015 0.9 0.060-0.006db 0.9 

WUF-W 0.051 0.020 0.9 0.064 0.9 

FF 0.061-0.00064db 0.020 0.9 0.080-0.00064db 0.9 

RBS 0.060-0.0003 db 0.020 0.9 0.080-0.0003 db 0.9 

WFP 0.04 0.020 0.9 0.07 0.9 

BUEP 0.071-0.0013 db .015 0.9 0.081-0.0013 db 0.9 

BSEP 0.071-0.0013 db .015 0.9 0.081-0.0013 db 0.9 

BFP 0.12-0.002 db .015 0.9 0.10-0.001 db 0.9 

DST 0.12-0.0032 db .015 0.9 0.14-0.0032 db 0.9 

Note: db is the beam depth, inches 

The capacity values indicated in Table 4-14 corresponding to collapse prevention 

behavior, have been conservatively established, based generally on engineering 

judgement, as few of the laboratory tests conducted actually loaded the connections to 

the point where failure to provide gravity-load resistance occurred. Accordingly, these 

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Recommended Seismic Design 

Chapter 4: Performance Evaluation Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

values should be used with caution. For more information refer to FEMA-355D State of 

the Art Report on Connection Performance. 

Tables 4-12 and 4-14 also present data on a drift angle capacity parameter 

indicated as qSD. This is the drift angle at which strength degradation initiates in 

the connection assembly, which may be controlled by connection degradation (for 

example, fracture of a weld or bolt) or framing behavior (such as buckling of the 

beam). The value of qSD is not directly used as a capacity index for evaluating 

performance. However, it is a critical parameter for determining whether a 

connection qualifies as a Special or Ordinary Moment Frame. Refer to Chapter 3 

and Appendix A for additional information on this topic. 

Table 4-13 Uncertainty Coefficient bUT for Local Interstory Drift Evaluation 

Building Height 

Performance Level 

Immediate Occupancy Collapse Prevention 

Special Moment Frames (SMF) 

Low Rise (3 stories or less) 0.30 0.30 

Mid Rise ( 4 – 12 stories) 0.30 0.35 

High Rise (> 12 stories) 0.30 0.40 

Ordinary Moment Frames (OMF) 

Low Rise (3 stories or less) 0.30 0.35 

Mid Rise ( 4 – 12 stories) 0.30 0.40 

High Rise (> 12 stories) 0.30 0.40 

Notes: 1. Value of bUT should be increased by 0.05 for linear static analysis 

2. Value of bUT may be reduced by 0.05 for nonlinear dynamic analysis 

Table 4-14 Behavior States for Performance Evaluation of Connection Assemblies 

Drift Angle Performance Level Description 

qIO Immediate Occupancy The lowest drift angle at which any of behaviors 1, 2, or 3, above, 

occur. 

qCP Collapse Prevention The drift angle at which behavior 4 occurs 

qSD - The lowest drift angle at which any of behaviors 2, 3, or 4 occur 

Note: The description involves the behavior list in the Commentary to Section 4.6.2.2.2. 

4.6.3 Performance Limited by Column Compressive Capacity 

4.6.3.1 Column Compressive Demand 

The factored column compressive demand shall be determined for each column as the 

quantity g gaD, where: 

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g

b

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Moment-Frame Buildings Chapter 4: Performance Evaluation

Criteria for New Steel FEMA-350 

D = the compressive axial load on the column determined as the sum of the Dead Load, 

25% of the unreduced Live Load, and the Seismic Demand. Seismic demand shall be 

determined by either of the following four analysis methods: 

Linear: The axial demands may be those predicted by a linear static or 

linear dynamic analysis, conducted in accordance with Section 4 of 

these Recommended Criteria. 

Plastic: The axial seismic demands may be based on plastic analysis, as 

indicated in Equation 4-5 of Section 4.4.3.3.6 of these 

Recommended Criteria. 

Nonlinear Static: The axial demands may be based on the computed forces from a 

nonlinear static analysis, at the target displacement, in accordance 

with Section 4 of these Recommended Criteria. 

Nonlinear Dynamic: The axial demands may be based on the computed design forces 

from a nonlinear dynamic analysis, in accordance with Section 4 of 

these Recommended Criteria. 

ga = Analysis uncertainty factor, taken from Table 4-15. 

g = Demand variability factor, taken as having the constant value 1.05. 

The uncertainty coefficient bUT shall be taken as indicated in Table 4-15 based on the 

procedure used to calculate column compressive demand D. 

Table 4-15 Analysis Uncertainty Factor ga and Total Uncertainty Coefficient bUT for 

Evaluation of Column Compressive Demands 

Analytical Procedure Analysis Uncertainty 

Factor 

ga 

Total Uncertainty 

Coefficient bUT 

Linear Static or Dynamic Analysis 1.15 0.35 

Plastic Analysis (Section 4.4.3.3.6) 1.0 0.15 

Nonlinear Static Analysis 1.05 0.20 

Nonlinear Dynamic Analysis 2

1.4 

e 

b 2

0.0225 + b

Note: b may be taken as the coefficient of variation (COV) of the axial load values determined from the suite of 

nonlinear analyses. 

Commentary: The value of g has been computed assuming a coefficient of 

variation for axial load values resulting from material strength variation and 

uncertainty in dead and live loads of 15%. The values of ga have been calculated 

assuming coefficients of variation of 30%, 0% and 15% related to uncertainty in 

the analysis procedures for linear, plastic and nonlinear static analyses, 

respectively. In reality, for structures that are stressed into the inelastic range, 

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Chapter 4: Performance Evaluation Moment-Frame Buildings 

FEMA-350 Criteria for New Steel 

elastic analysis will typically overestimate axial column demands, in which case, 

a value of 1.0 could be used for ga (0% coefficient of variation). However, for 

structures that are not loaded into the inelastic range, the indicated value is 

appropriate. Plastic analysis will also typically result in an upper-bound 

estimate of column demand, and application of additional demand factors is not 

appropriate. For nonlinear dynamic analysis, using a suite of ground motions, 

direct calculation of the analysis demand factor is possible, using the equation 

shown. All of these demand factors are based on the hazard parameter k having a 

value of 3.0, typical of conditions in coastal California. 

4.6.3.2 Column Compressive Capacity 

Factored compressive capacity of each column to resist compressive axial loads, shall be 

determined as the product of the resistance factor f and the nominal axial strength C of the 

column, which shall be determined in accordance with the AISC Load and Resistance Factor 

Design Specification. For the purposes of this evaluation, the effective length coefficient k shall 

be taken as having a value of 1.0 and the resistance factor f shall be assigned a value of 0.95. 

4.6.4 Column Splice Capacity 

The capacity of column tensile splices, other than splices consisting of complete joint 

penetration (CJP) groove welds of all elements of the column (flanges and webs) shall be 

evaluated in accordance with this section. Column splices consisting of CJP welds of all 

elements of the column, and in which the weld filler metal conforms to the recommendations of 

Section 3.3.2.4 of these Recommended Criteria need not be evaluated. 

Commentary: Welded splices in which the flanges and welds of the butting 

sections are joined with CJP groove welds and in which weld access holes are 

provided in accordance with AISC requirements qualify as having all elements 

joined by CJP groove welds. 

4.6.4.1 Column Splice Tensile Demand 

Factored column splice tensile demand shall be determined for each column as the quantity 

g gaD where D is the column splice tensile demand. Column splice tensile demand shall be 

determined as the computed Seismic Demand in the column less 90% of the computed Dead 

Load demand. Seismic Demand shall be as determined for column compressive demand, in 

accordance with Section 3.6.3.1. The demand variability factor g shall be taken as having a value 

of 1.05 and the analysis uncertainty factor ga shall be taken as indicated in Table 4-15. The total 

uncertainty coefficient bUT shall also be taken as indicated in Table 4-15. 

4.6.4.2 Column Splice Tensile Capacity 

The capacity of individual column splices to resist tensile axial loads shall be determined as 

the product of the resistance factor f and the nominal tensile strength C of the splice, as 

determined in accordance with the AISC Load and Resistance Factor Design Specification. 

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Moment-Frame Buildings Chapter 4: Performance Evaluation

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Specifically, Chapter J therein shall be used to calculate the nominal tensile strength of the splice 

connection. For the purposes of this evaluation, f shall be assigned a value of 0.9. 

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Criteria for New Steel FEMA-350

Moment-Frame Buildings Appendix A: Detailed Procedures for Performance Evaluation

A. DETAILED PROCEDURES FOR PERFORMANCE EVALUATION 

A.1 Scope 

This appendix provides detailed procedures for evaluating the performance capability of steel 

moment-frame buildings. These detailed procedures are provided as a supplement to the 

simplified performance evaluation procedures in Chapter 4. They may be used to demonstrate 

enhanced levels of confidence with regard to the ability of a particular building to meet desired 

performance objectives, relative to the confidence levels that may be derived using the more 

simplified procedures, and they must be used instead of the procedures of Chapter 4, for irregular 

structures and for structures with connections that have not been prequalified. This appendix 

also provides criteria for performance evaluation for deterministically defined hazards. 

Commentary: Chapter 4 provides procedures for a simplified method of 

performance evaluation, using factored-demand-to-capacity ratios to determine a 

level of confidence with regard to a building’s ability to provide a desired 

performance objective. The tabular values of demand and resistance factors and 

confidence indices contained in Chapter 4 were derived using the procedures 

presented in this appendix, applied to the performance evaluation of a suite of 

regularly configured model buildings. Since this suite of model buildings is not 

completely representative of any individual structure, the use of the tabular values 

inherently entails some uncertainty, and thus reduced levels of confidence, with 

regard to performance prediction. The detailed procedures in this appendix 

permit reduction in these uncertainties, and therefore enhanced confidence, with 

regard to prediction of building performance. These more detailed procedures 

must be used for those irregular building configurations not well represented by 

the model buildings used as the basis for the values contained in Chapter 4. 

A.2 Performance Evaluation Approach 

A.2.1 Performance Objectives and Confidence 

As defined in Section 4.2 of these Recommended Criteria, performance is defined in terms of 

probabilistic performance objectives. A performance objective consists of the specification of a 

performance level and an acceptable low probability that poorer performance could occur within 

a specific period of time, typically taken as 50 years. Alternatively, deterministic performance 

objectives can also be evaluated. Deterministic performance objectives consist of the 

specification of a performance level and a specific earthquake, that is, fault location and 

magnitude, for which this performance is to be attained. 

Two performance levels are defined: the Immediate Occupancy performance level and the 

Collapse Prevention performance level. Detailed descriptions of these performance levels may 

be found in Chapter 4. The evaluation procedures contained in this appendix permit estimation 

of a level of confidence associated with achievement of a performance objective. For example, a 

design may be determined to provide a 95% level of confidence that there is less than a 2% 

probability in 50 years of more severe damage than represented by the Collapse Prevention level. 

A-1
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

Appendix A: Detailed Procedures for Performance Evaluation Moment-Frame Buildings 

For another example, a design may be determined to provide a 50% level of confidence that the 

structure will provide Immediate Occupancy performance, or a better performance, for a Richter 

magnitude 6 earthquake along a defined fault. 

Commentary: The probability that a building may experience damage more 

severe than that defined for a given performance level is a function of two 

principal factors. The first of these is the structure’s vulnerability, that is, the 

probability that it will experience certain levels of damage given that it 

experiences ground motion of certain intensity. The second of these factors is the 

site hazard, that is, the probability that ground shaking of varying intensities may 

occur in a given time period. The probability that damage exceeding a given 

performance level may occur in a period of time is calculated as the integral over 

time of the probability that damage will exceed that permitted within a 

performance level. Mathematically, this may be expressed as: 

P(D > PL) = � PD>PL (x)h(x)dx (A-1) 

where: 

P(D>PL) = Probability of damage exceeding a performance level in a period 

of t years 

PD>PL(x) = Probability of damage exceeding a performance level given that 

the ground motion intensity is level x, as a function of x, 

h(x)dx = probability of experiencing a ground motion intensity of level (x) to 

(x + dx) in a period of t years 

Vulnerability may be thought of as the capacity of the structure to resist greater 

damage than that defining a performance level. Structural response parameters 

that may be used to measure capacity include the structure’s ability to undergo 

global building drift, maximum tolerable member forces, and maximum tolerable 

inelastic deformations. Ground accelerations associated with the seismic hazard, 

and the resulting enforced global building drift, member forces and inelastic 

deformations produced by the hazard may be thought of as demands. If both the 

demand that a structure will experience over a period of time and the structure’s 

capacity to resist this demand could be perfectly defined, then performance 

objectives, the probability that damage may exceed a performance level within a 

period of time, could be ascertained with 100% confidence. However, the process 

of predicting the capacity of a structure to resist ground shaking demands as well 

as the process of predicting the severity of demands that will actually be 

experienced entail significant uncertainties. Confidence level is a measure of the 

extent of uncertainty inherent in this process. A level of 100% confidence may be 

described as perfect confidence. In reality, it is never possible to attain such 

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confidence. Confidence levels on the order of 90 or 95% are considered high, 

while confidence levels less than 50% are considered low. 

Generally, uncertainty can be reduced, and confidence increased, by 

obtaining better knowledge or using better procedures. For example, enhanced 

understanding and reduced uncertainty with regard to the prediction of the effects 

of ground shaking on a structure can be obtained by using a more accurate 

analytical procedure to predict the structure’s response. Enhanced 

understanding of the capacity of a structure to resist ground shaking demands can 

be obtained by obtaining specific laboratory data on the physical properties of the 

materials of construction and on the damageability of individual beam-column 

connection assemblies. 

The simplified performance evaluation procedures of Chapter 4 are based on 

the typical characteristics of standard buildings. Consequently, they incorporate 

significant uncertainty in the performance prediction process. As a result of this 

significant uncertainty, it is anticipated that the actual ability of a structure to 

achieve a given performance objective may be significantly better than would be 

indicated by those simple procedures. The more detailed procedures of this 

appendix may be used to improve the definition of the actual uncertainties 

incorporated in the prediction of performance for a specific structure and thereby 

to obtain better confidence with regard to the prediction of performance for an 

individual structure. 

As an example, using the simplified procedures of Chapter 4, it may be found 

that for a specific structure, there is only a 50% level of confidence that there is 

less than a 10% chance in 50 years of poorer performance than the Collapse 

Prevention level. This rather low level of confidence may be more a function of 

the uncertainty inherent in the simplified procedures than the actual inadequate 

capacity of the building to provide Collapse Prevention performance. In such a 

case, it may be possible to use the procedures contained in this appendix to 

reduce the uncertainty inherent in the performance estimation and find that 

instead, there may be as much as a 95% level of confidence in obtaining such 

performance. 

In both the procedures of this appendix and Chapter 4, the uncertainties 

associated with estimation of the intensity of ground motion have been neglected. 

These uncertainties can be quite high, on the order of those associated with 

structural performance or even higher. Thus, the confidence estimated using 

these procedures is really a confidence with regard to structural performance, 

given the seismicity as portrayed by the USGS hazard maps that accompany 

FEMA-273 and FEMA-302. 

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A.2.2 Basic Procedure 

As indicated in Chapter 4, a demand and resistance factor design (DRFD) format is used to 

associate a level of confidence with the probability that a building will have less than a specified 

probability of exceedance of a desired performance level. The basic approach is to determine a 

confidence parameter, l, which may then be used, with reference to Table A-1, to determine the 

confidence level that exists with regard to performance estimation. The confidence parameter, l, 

is determined from the factored-demand-to-capacity equation: 

l = gg aD (A-2) fC 

where: 

C = median estimate of the capacity of the structure. This estimate may be obtained either 

by reference to default values contained in Chapter 4, or by more rigorous direct 

calculation of capacity using the procedures of this appendix, 

D = calculated demand on the structure, obtained from a structural analysis, 

g = a demand variability factor that accounts for the variability inherent in the prediction 

of demand related to assumptions made in structural modeling and prediction of the 

character of ground shaking, 

ga = an analysis uncertainty factor that accounts for the bias and uncertainty associated 

with the specific analytical procedure used to estimate structural demand as a function 

of ground shaking intensity, 

f = a resistance factor that accounts for the uncertainty and variability inherent in the 

prediction of structural capacity as a function of ground shaking intensity, 

l = a confidence index parameter from which a level of confidence can be obtained by 

reference to Table A-1. 

Several structural response parameters are used to evaluate structural performance. The 

primary parameter used for this purpose is interstory drift. Interstory drift is an excellent 

parameter for judging the ability of a structure to resist P-D instability and collapse. It is also 

closely related to plastic rotation demand, or drift angle demand, on individual beam-column 

connection assemblies, and therefore a good predictor of the performance of beams, columns and 

connections. Other parameters used in these guidelines include column axial compression and 

column axial tension. In order to determine a level of confidence with regard to the probability 

that a building has less than a specified probability of exceeding a performance level over a 

period of time, the following steps are followed: 

1. The performance objective to be evaluated is selected. This requires selection of a 

performance level of interest, for example, Collapse Prevention or Immediate Occupancy, 

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and a desired probability that damage in a period of time will be worse than this performance 

level. Representative performance objectives may include: 

• 2% probability of poorer performance than Collapse Prevention level in 50 years 

• 50% probability of poorer performance than Immediate Occupancy level in 50 years. 

It is also possible to express performance objectives in a deterministic manner, where 

attainment of the performance is conditioned on the occurrence of a specific magnitude 

earthquake on an identified fault. 

2. Characteristic motion for the performance objective is determined. For probabilistic 

performance objectives, an average estimate of the ground shaking intensity at the 

probability of exceedance identified in the performance objective definition (step 1) is 

determined. For example, if the performance objective is a 2% probability of poorer 

performance than Collapse Prevention level in 50 years, then an average estimate of ground 

shaking demands with a 2% probability of exceedance in 50 years would be determined. 

Ground shaking intensity is characterized by the parameter SaT1, the 5% damped spectral 

response acceleration at the site for the fundamental period of response of the structure. 

FEMA-273 provides procedures for determining this parameter for any probability of 

exceedance in a 50-year period. 

For deterministic performance objectives, an average estimate of the ground motion at the 

building site for the specific earthquake magnitude and fault location must be made. As with 

probabilistic estimates, the motion is characterized by SaT1. 

3. Structural demands for the characteristic earthquake ground motion are determined. 

A mathematical structural model is developed to represent the building structure. This model 

is then subjected to a structural analysis, using any of the methods contained in Chapter 4. 

This analysis provides estimates of maximum interstory drift demand, maximum column 

compressive demand, and maximum column-splice tensile demand, for the ground motion 

determined in step 2. 

4. Median estimates of structural capacity are determined. Median estimates of the 

interstory drift capacity of the moment-resisting connections and the building frame as a 

whole are determined, as are median estimates of column compressive capacity and columnsplice tensile capacity. Interstory drift capacity for the building frame, as a whole, may be 

estimated using the default values of Chapter 4 for regular structures, or alternatively, the 

detailed procedures of Section A.6 may be used. These detailed procedures are mandatory 

for irregular structures. Interstory drift capacity for moment-resisting connections that are 

prequalified in Chapter 3 of these Recommended Criteria may be estimated using the default 

values of Chapter 4, or alternatively, direct laboratory data on beam-column connection 

assembly performance capability and the procedures of Section A.5 of this appendix may be 

used. Median estimates of column compressive capacity and column-splice tensile capacity 

are made using the procedures of Chapter 4. 

5. A factored-demand-to-capacity ratio, l is determined. For each of the performance 

parameters, i.e., interstory drift as related to global building frame performance, interstory 

drift as related to connection performance, column compression, and column splice tension, 

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Equation A-2 is independently applied to determine the value of the confidence parameter l. 

In each case, the calculated estimates of demand D and capacity C are determined using steps 

3 and 4, respectively. If the procedures of Chapter 4 are used to determine either demand or 

median capacity estimates, then the corresponding values of the demand factors g and 

resistance factors f should also be determined in accordance with the procedures of Chapter 

4. If the procedures of this appendix are used to determine median demand, or capacity, then 

the corresponding demand and resistance factors should be determined in accordance with 

the applicable procedures of this appendix. 

6. Evaluate confidence. The confidence obtained with regard to the ability of the structure to 

meet the performance objective is determined using the lowest of the l values determined in 

accordance with step 5 above, back-calculated from the equation: 

l = e 

-bbUT ( KX -k bUT 2) (A-3) 

where: 

b = a coefficient relating the incremental change in demand (drift, force, or 

deformation) to an incremental change in ground shaking intensity, at the 

hazard level of interest, typically taken as having a value of 1.0, 

bUT = an uncertainty measure equal to the vector sum of the logarithmic standard 

deviation of the variations in demand and capacity resulting from uncertainty, 

k = the slope of the hazard curve, in ln-ln coordinates, at the hazard level of 

interest, i.e., the ratio of incremental change in SaT1 to incremental change in 

annual probability of exceedance (refer to Section A.3.2), 

KX = standard Gaussian variate associated with probability x of not being exceeded 

as a function of number of standard deviations above or below the mean found 

in standard probability tables. 

Table A-1 provides a solution for this equation, for various values of the parameters, 

k, l, and bUT. 

The values of the parameter bUT used in Equation A-3 and Table A-1 are used to account for 

the uncertainties inherent in the estimation of demands and capacities. Uncertainty enters the 

process through a variety of assumptions that are made in the performance evaluation process, 

including, for example, assumed values of damping, structural period, properties used in 

structural modeling, and strengths of materials. Assuming that the amount of uncertainty 

introduced by each of the assumptions can be characterized, the parameter bUT can be calculated 

using the equation: 

bUT = 2

ui �i 

b (A-4) 

where: bui are the standard deviations of the natural logarithms of the variation in demand or 

capacity resulting from each of these various sources of uncertainty. Sections A.4, A.5 and A.6 

indicate how to determine bui values associated with demand estimation, beam-column 

connection assembly behavior, and building global stability capacity prediction, respectively. 

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A.3 Determination of Hazard Parameters 

Two basic hazard parameters are required by these performance evaluation procedures. The 

first of these, SaT1, is the median, 5%-damped, linear spectral response acceleration, at the 

fundamental period of the building, at the desired hazard level (probability of exceedance in a 

50-year period or specific earthquake magnitude and fault). Section A.3.1 provides guidelines 

for obtaining this parameter. The second parameter is the slope k of the hazard curve in 

logarithmic space, also evaluated at the desired hazard level. Section A.3.2 provides guidelines 

for obtaining this parameter. 

A.3.1 Spectral Response Acceleration 

Probabilistic, 5%-damped, linear spectral response acceleration, SaT1 at the fundamental 

period of the building, at the desired hazard level (probability of exceedance in a 50-year period), 

may be determined in several different ways. These include: 

a. Site-specific seismological and geotechnical investigation. FEMA-273 provides guidelines 

for this method. 

b. Use of national hazard maps developed by the United States Geologic Survey. FEMA-273 

also provides guidelines for the use of these maps for this purpose. 

Deterministic 5%-damped, linear spectral response acceleration SaT1 at the fundamental 

period of the building shall be determined based on site-specific seismological and geologic 

study. 

The spectral response acceleration SaT1 is used as a reference point, through which a response 

spectrum is plotted. This response spectrum may be used directly in the structural analysis, or 

alternatively, may be used as a basis for the development of ground motion accelerograms used 

in the structural analysis. Refer to Chapter 4 for guidelines on analysis. 

A.3.2 Logarithmic Hazard Curve Slope 

In these procedures, the logarithmic slope k of the hazard curve at the desired hazard level is 

used to determine the resistance factors, demand factors and also the confidence levels. The 

hazard curve is a plot of probability of exceedance of a spectral amplitude versus that spectral 

amplitude, for a given period, and is usually plotted on a log-log scale. In functional form it can 

be represented by the equation: 

H Si (Si ) = k0 Si 

-k (A-5) 

where: 

HSi(Si) = the probability of ground shaking having a spectral response acceleration 

greater than Si, 

k0 = a constant, dependent on the seismicity of the individual site, 

k = the logarithmic slope of the hazard curve. 

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The slope of the hazard curve is a function of the hazard level, location and response period. 

USGS maps provide values of 5%-damped, spectral response accelerations at periods of 0.2 

seconds, termed Ss, and 1 second, termed S1, for ground motions having 2% and 10% 

probabilities of exceedance in 50 years, for all locations in the United States. This information is 

also available on their web site and on a CD-ROM. Since most steel moment-frames have 

relatively long fundamental periods, the slope of the hazard curve may be determined for most 

such structures using the S1 values published by the USGS for probabilities of exceedance of 2% 

and 10% in 50 years, and substitution of these values into the following equation: 

� H � ln �

� H

S1(10/50) � 

k = Ł S1(2/50) ł = 

1.65 (A-6) � S � � S � ln � 1(2/50) � ln � 1(2/50) � � S1(10/50) ł

� Ł� S1(10/50) ł� Ł 

where: 

S1(10/50) = spectral amplitude for 10/50 hazard level 

S1(2/50) = spectral amplitude for 2/50 hazard level 

HS1(10/50) = probability of exceedance for 10% in 50 years = 1/475 = 0.0021 

HS1(2/50) = probability of exceedance for 2% in 50 years = 1/2475 = 0.00040 

The accompanying sidebar provides an example of how k may be determined using this 

procedure, for a representative site. As an alternative to using this detailed procedure, an 

approximate value of k may be obtained from Table A-2. When deterministic ground shaking 

demands (specific magnitude earthquake on a fault) are used as the basis for a performance 

objective, the value of k shall be taken as 4.0, regardless of the site seismicity. 

Table A-2 Default Values of the Logarithmic Hazard Curve Slope k 

for Probabilisitc Ground Shaking Hazards 

Region k 

Alaska, California and the Pacific 

Northwest 

3 

Intermountain Region, Basin & 

Range Tectonic Province 

2 

Other U.S. locations 1 

Note: For deterministic ground shaking demands, use a value of k = 4.0 

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Example determination of the parameter, k, the logarithmic slope of the hazard curve using 

hazard data from the USGS. 

Example site location: 

Referencing USGS maps, web site, find S1(10/50)

 = 0.45g, S1(2/50) = 0.77g 

Substituting into equation A-5, find: 

3 .07

0 .537

1 .65

0 .45

0.77 ln 

1 .65

= =

� 

ł 

� � 

Ł 

� = 

g 

g 

k 

Los Angeles City Hall 

A.4 Determination of Demand Factors 

The demand variability factor g and analysis uncertainty factor ga are used to adjust the 

calculated interstory drift, column axial load and column-splice tension demands to their mean 

values, considering the variability and uncertainty inherent in drift demand prediction. 

Variability in drift demand prediction is primarily a result of the fact that due to relatively 

subtle differences in acceleration records, a structure will respond somewhat differently to 

different ground motion records, even if they are well characterized by the same response 

spectrum. Since it is not possible to predict the exact acceleration record that a structure may 

experience, it is necessary to account for the probable variation in demand produced by all 

possible different records. This is accomplished by developing a nonlinear mathematical model 

of the structure, and running nonlinear response history analyses of the structure for a suite of 

ground motion records, all of which are scaled to match the 5% damped linear spectral response 

acceleration, SaT1, described in Section A.3.1. From these analyses, statistics are developed for 

the median value and standard deviation of the natural logarithm of the various demand 

parameters including maximum interstory drift, column axial load, and column splice tension. 

These standard deviations of the natural logarithms of these response parameters are denoted 

bDR. 

Once the value of bDR has been determined, the demand variability factor, g, is calculated 

from the equation: 

k 2 

g = e 2b 

bDR (A-7) 

where: 

k is the logarithmic slope of the hazard curve, taken in accordance with 

Section A.3.1 

b is a coefficient that represents the amount that demand increases as a 

function of hazard, and may normally be taken as having a value of 1.0 

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Uncertainty in the prediction of demands is due to an inability to define accurately the value 

of such parameters as the yield strength of the material, the viscous damping of the structure, the 

effect of nonstructural components, the effect of foundation flexibility on overall structural 

response, and similar modeling issues. Although it is not feasibly practical to do so, it is 

theoretically possible to measure each of these quantities for a building and to model their effects 

exactly. Since it is not practical to do this, instead likely values are used for each of these effects 

in the model, to account for the possible inaccuracies introduced by using these likely values, 

rather than real values. These inaccuracies are accounted for by developing a series of models to 

represent the structure, accounting for the likely distribution of these various parameters. Each 

of these models is used to run analyses with a single ground motion record, and statistics are 

developed for the effect of variation in these parameters on predicted demands. As with the 

variability due to ground motion, the standard deviation of the natural logarithms of the response 

parameters are calculated, and denoted by bDU. This parameter is used to calculate the analytical 

uncertainty factor, ga. 

In addition to uncertainty in demand prediction, the analytical uncertainty factor ga also 

accounts for inherent bias, that is, systematic under- or over-prediction of demand, inherent in an 

analytical methodology. Bias is determined by using the analytical methodology, for example, 

elastic modal analysis, to predict demand for a suite of ground motions and then evaluating the 

ratio of the demand predicted by nonlinear time history analysis of the structure to that predicted 

by the methodology for the same ground motion. This may be represented mathematically as: 

CB = 

demand predicted by nonlinear time history analysis (A-8) demand predicted by analysis method 

where CB is the bias factor. The bias factor that is applicable to a specific structure is taken as 

the median value of CB calculated from a suite of ground motions. The variation in the bias 

factors obtained from this suite of ground motions is used as one of the components in the 

calculation of bDU. 

Once the median bias factor, CB and logarithmic standard deviation in demand prediction bDU 

have been determined, the analysis uncertainty factor, ga is calculated from the equation: 

k 2 bDU

g a = C e 2b (A-9) B 

The analysis uncertainty factors presented in Chapter 3 were calculated using this approach 

as applied to a suite of typical buildings. In addition to the uncertainties calculated using this 

procedure, it was assumed that even the most sophisticated methods of nonlinear time history 

analysis entail some uncertainty relative to the actual behavior of a real structure. Additional 

uncertainty was associated with other analysis methods to account for effects of structural 

irregularity, which were not adequately represented in the suite of model buildings used in the 

study. The value of the total logarithmic uncertainty bDU used as a basis for the analysis 

uncertainty factors presented in Chapter 4 are summarized in Table A-3. The bias factors CB 

used in Chapter 4 are summarized in Table A-4. It is recommended that these default values for 

CB and bDU be used for all buildings. If it is desired to calculate building-specific bDU values, it 

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is recommended that these values not be taken as less than those indicated in Table A-3 for 

nonlinear dynamic analysis, for the applicable building characteristics. 

Table A-3 Default Logarithmic Uncertainty bDU for Various Analysis Methods 

Analysis Procedure 

Linear Static Linear 

Dynamic 

Nonlinear 

Static 

Nonlinear 

Dynamic 

Performance Level IO CP IO CP IO CP IO CP 

Type 1 Connections 

Low Rise (<4 stories) 0.17 0.22 0.15 0.16 0.14 0.17 0.10 0.15 

Mid Rise (4 – 12 stories) 0.18 0.29 0.15 0.23 0.15 0.23 0.13 0.20 

High Rise (> 12 stories) 0.31 0.25 0.19 0.29 0.17 0.27 0.17 0.25 

Type 2 Connections 

Low Rise (<4 stories) 0.19 0.23 0.16 0.25 0.18 0.18 0.10 0.15 

Mid Rise (4 – 12 stories) 0.20 0.30 0.17 0.33 0.14 0.21 0.13 0.20 

High Rise (> 12 stories) 0.21 0.36 0.21 0.31 0.18 0.33 0.17 0.25 

Table A-4 Default Bias Factors CB 

Analysis Procedure 

Linear Static Linear 

Dynamic 

Nonlinear 

Static 

Nonlinear 

Dynamic 

Performance Level IO CP IO CP IO CP IO CP 

Type 1 Connections 

Low Rise (<4 stories) 0.90 0.65 1.00 0.80 1.10 0.85 1.00 1.00 

Mid Rise (4 – 12 stories) 1.10 0.85 1.10 1.15 1.40 0.95 1.00 1.00 

High Rise (> 12 stories) 1.05 1.0 1.15 1.0 1.30 0.85 1.00 1.00 

Type 2 Connections 

Low Rise (<4 stories) 0.75 0.90 1.00 1.20 0.90 1.25 1.00 1.00 

Mid Rise (4 – 12 stories) 0.80 1.00 1.05 1.30 1.08 1.35 1.00 1.00 

High Rise (> 12 stories) 0.75 0.70 1.30 1.20 1.30 1.30 1.00 1.00 

Commentary: Although it may be possible, for certain structures, to increase the 

confidence associated with a prediction of probable earthquake demands on the 

structure, through calculation of structure-specific analysis uncertainty factors, in 

general this is a very laborious process. It is recommended that the default 

values of bDU and CB, contained in Tables A-3 and A-4, be used for most 

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structures. However, the procedures of this section can be used to adjust the 

analysis uncertainty and demand variability factors for the site seismicity k. 

A.5 Determination of Beam-Column Connection Assembly Capacities 

The probable behavior of beam-column connection assemblies at various demand levels can 

best be determined by full-scale laboratory testing. Such testing can provide indications of the 

probable physical behavior of such assemblies in buildings. Depending on the characteristics of 

the assembly being tested, meaningful behaviors may include the following: onset of local 

buckling of flanges; initiation of fractures in welds, base metal or bolts; a drop in the moment 

developed by the connection beyond predetermined levels; or complete failure, at which point 

the connection is no longer able to maintain attachment between the beam and column under the 

influence of gravity loads. If sufficient laboratory data are available, it should be possible to 

obtain statistics, including a median value and standard deviation, on the demand levels at which 

these various behaviors occur. 

In the past, most laboratories used plastic rotation as the demand parameter by which beamcolumn connection assembly behavior was judged. However, since plastic deformation may 

occur at a number of locations within a connection assembly, including within the beam itself, 

within the connection elements, and within the column panel zone or column, many laboratories 

have measured and reported plastic rotation angles from testing in an inconsistent manner. 

Therefore, in these Recommended Criteria, total interstory drift angle, as indicated in Section 

4.6, is the preferred demand parameter for reporting laboratory data. This parameter is less 

subject to erroneous interpretation by testing laboratories and also has the advantage that it is a 

quantity directly predicted by linear structural analyses. 

Median drift angle capacities, C, and resistance factors, f, for various prequalified 

connection types are presented in Chapter 4. These values were determined from cyclic tests of 

full-size connection assemblies using the testing protocols indicated in Section 3.9. The cyclic 

tests are used to determine the load-deformation hysteresis behavior of the system and the 

connection drift angle at which the following behaviors occur: 

1. onset of local flange buckling of beams, 

2. degradation of moment-resisting capacity of the assembly to a value below the nominal 

moment-resisting capacity, 

3. initiation of fracture of bolts, welds, or base metal that results in significant strength 

degradation of the assembly, and 

4. complete failure of the connection, characterized by an inability of the connection to 

maintain its integrity under gravity loading. 

Based on this data, drift angle statistics, including a median value and logarithmic standard 

deviation are obtained for the Immediate Occupancy and Collapse Prevention damage states, as 

indicated in Table A-5. The quantity qU, the ultimate capacity of the connection, is used to 

evaluate the acceptability of connection behavior for the Collapse Prevention performance level 

as limited by local behavior. 

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Table A-5 Behavior States for Performance Evaluation of Connection Assemblies 

Symbol Performance Level Description 

qIO Immediate Occupancy The lowest drift angle at which any of behaviors 1, 2, or 3, occur 

(see Section A.5, above) 

qU Ultimate The drift angle at which behavior 4 occurs 

qSD Strength Degradation The lowest drift angle at which any of behaviors 2, 3, or 4 occur 

A.5.1 Connection Test Protocols 

Two connection test protocols have been developed under this project. The standard protocol 

is intended to represent the energy input and cyclic deformation characteristics experienced by 

connection assemblies in steel moment frames which are subjected to strong ground shaking 

from large magnitude earthquakes, but which are not located within a few kilometers of the fault 

rupture. This protocol presented in Section 6.9 is similar to that contained in ATC-24 (ATC, 

1992) and consists of ramped cyclic loading, starting with initial cycles of low energy input 

within the elastic range of behavior of the assembly, and progressing to increasing deformation 

of the beam tip until assembly failure occurs. However, unlike ATC-24, the protocol 

incorporates fewer cycles of large-displacement testing to balance more closely the energy input 

to the assembly, with that likely experienced by framing in a real building. The second protocol 

is intended to represent the demands experienced by connection assemblies in typical steel 

moment-frame buildings responding to near-fault ground motion, dominated by large velocity 

pulses. This protocol (Krawinkler et al., 2000) consists of an initial single large displacement, 

representing the initial response of a structure to a velocity pulse, followed by repeated cycles of 

lesser displacement. 

Performance characteristics of connection assemblies, for use in performance evaluation of 

buildings, should be selected based on the characteristics of earthquakes dominating the hazard 

for the building site, at the specific hazard level. Most buildings are not located on sites that are 

likely to be subjected to ground shaking with near-field pulse characteristics. Connection 

performance data for such buildings should be based on the standard protocols. Buildings on 

sites that are close to a major active fault are most likely to experience ground shaking with these 

strong pulse-like characteristics and connection performance for such buildings should be based 

on the near-fault protocol. However, qualification of connections for classification as either 

Special Moment Frame or Ordinary Moment Frame connections should be based on the standard 

protocol. 

A.5.2 Determination of Beam-Column Assembly Capacities and Resistance Factors 

Median drift angle capacities for the quantities qIO and qU should be taken directly from 

available laboratory data. The median value should be taken as that value from all of the 

available tests that is not exceeded by 50% of the tests. The value of the quantity f, for each of 

the Immediate Occupancy and ultimate (Collapse Prevention) states should be determined by the 

following procedure. 

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1. Obtain the logarithmic standard deviation of the qIO or qU values available from the 

laboratory data. That is, take the standard deviation of the natural logarithms of the qIO or qU 

values respectively, obtained from each laboratory test. Logarithmic standard deviation may 

be determined from the formula: 

b = 

( )

1 

ln ln 

1 

2 

-

� - =

n 

x x n 

i i i (A-10) 

where: 

b = the standard deviation of the natural logarithms of the test data

xi = individual test data value

n = the number of tests from which data is available

ln xi =the mean of the logarithms of the xi values. 

2. Calculate the connection resistance factor fR due to randomness, the observed variation in 

connection behavior, from laboratory testing, using the equation: 

k b 2 -

fR = e 2b (A-11) 

where:

k = the slope of the hazard curve, determined in accordance with Section A.3.2

b = a coefficient that relates the change in hazard to the change in demand, and which

may be taken as having a value of 1.0 

b = the logarithmic standard deviation calculated in accordance with Equation A-10. 

3. Determine the connection resistance factor accounting for random and uncertain behaviors 

from the equation: 

- k (0.2)

2 

f = fRfU = fRe 2b (A-12) 

where:

fR = the resistance factor accounting for random behavior

fU = the resistance factor accounting for uncertainty in the relationship between

laboratory findings and behavior in real buildings, and assumed in these 

Recommended Criteria to have a logarithmic standard deviation bu of 0.2 

A.6 Global Stability Capacity 

For the Collapse Prevention performance level, in addition to consideration of local behavior, 

that is, the damage sustained by individual beams and beam-column connection assemblies, it is 

also important to consider the global stability of the frame. The procedures indicated in this 

section are recommended for determining an interstory drift capacity C and resistance factor f 

associated with global stability of the structure. 

A-15
Recommended Seismic Evaluation and Upgrade 

FEMA-350 Criteria for Existing Welded 

Appendix A: Detailed Procedures for Performance Evaluation Steel Moment-Frame Buildings 

The global stability limit is determined using the Incremental Dynamic Analysis (IDA) 

technique. This requires the following steps: 

1. Choose a suite of ten to twenty accelerograms representative of the site and hazard level for 

which the Collapse Prevention level is desired to be achieved. 

2. Select one of these accelerograms and perform an elastic time-history analysis of the 

building. Determine a scaling factor for this accelerogram such that the elastic time history 

analysis would result in response that would produce incipient yielding in the structure. 

Determine the 5%-damped, spectral response acceleration SaT1 for this scaled accelerogram at 

the fundamental period of the structure. On a graph with an abscissa consisting of peak 

interstory drift and an ordinate axis of SaT1, plot the point consisting of the maximum 

calculated interstory drift from the scaled analysis and the scaled value of SaT1. Draw a 

straight line from the origin of the axes to this point. The slope of this line is referred to as 

the elastic slope, Se 

3. Increase the scaling of the accelerogram, such that it will produce mild nonlinear behavior of 

the building. Perform a nonlinear time-history analysis of the building for this scaled 

accelerogram. Determine the SaT1 for this scaled accelerogram and the maximum predicted 

interstory drift from the analysis. Plot this point on the graph. Call this point D1. 

4. Increase the scaling amplitude of the accelerogram slightly and repeat Step 3. Plot this point 

as D2. Draw a straight line between points D1 and D2. 

5. Repeat Step 4 until the straight line slope between consecutive points Di and Di+1, is less than 

0.2 Se. When this condition is reached, Di+1 is the global drift capacity for this accelerogram. 

If Di+1 > 0.10 then the drift capacity is taken as 0.10. Figure A-1 presents a typical series of 

plots obtained from such analyses. 

6. Repeat Steps 2 through 5 for each of the accelerograms in the suite selected as representative 

of the site and hazard and determine an interstory drift capacity for the structure for each 

accelerogram. 

7. Determine a median interstory drift capacity C for global collapse as the median value of the 

calculated set of interstory drift capacities, determined for each of the accelerograms. The 

median value is that value exceeded by 50% of the accelerograms. 

8. Determine a logarithmic standard deviation b for random differences in ground motion 

accelerograms, using Equation A-10 of Section A.5.2. In this equation, xi is the interstory 

drift capacity predicted for the i th accelerogram, and n is the number of accelerograms 

contained in the analyzed suite. 

9. Calculate the global resistance factor fR due to randomness in the predicted global collapse 

capacity for various ground motions from the equation: 

k b 2 -

fR = e 2b (A-13) 

where k and b are the parameters described in Section A.5.2 and b is the logarithmic standard 

deviation calculated in the previous step. 

A-16
Recommended Seismic Evaluation and Upgrade 

Appendix A: Detailed Procedures for Performance Evaluation 

Criteria for Existing Welded FEMA-350 

Steel Moment-Frame Buildings 

0 

0.5 

1 

1.5 

2 

2.5 

0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 

Max im um Interstory Drift 

Spectral Acceleration (at fundame ntal period), g 

LA23 

LA22 

LA30 

LA24 

LA28 

Figure A-1 Representative Incremental Dynamic Analysis Plots 

10. Determine a resistance factor for global collapse from the equation: 

k 2 - bU

f = fUfR = e 2b fR (A-14) 

where: 

fR is the global resistance factor due to randomness determined in Step 9. 

bU is the logarithmic standard deviation related to uncertainty in analytical prediction of 

global collapse prevention taken as having a value of 0.15 for low-rise structures, 3 

stories or less in height; a value of 0.2 for mid-rise structures, 4 stories to 12 stories in 

height; and taken as having a value of 0.25 for high-rise structures, greater than 12 stories 

in height. 

It is important that the analytical model used for determining the global drift demand be as 

accurate as possible. The model should include the elements of the moment-resisting frame as 

well as framing that is not intended to participate in lateral load resistance. A nominal viscous 

damping of 3% of critical is recommended for most buildings. The element models for beamcolumn assemblies should realistically account for the effects of panel zone flexibility and 

yielding, element strain hardening, and stiffness and strength degradation, so that the hysteretic 

behavior of the element models closely matches that obtained from laboratory testing of 

comparable assemblies. 

Commentary: As noted above, accurate representation of the hysteretic behavior 

of the beam-column assemblies is important. Earthquake-induced global collapse 

initiates when displacements produced by the response to ground shaking are 

A-17
Recommended Seismic Evaluation and Upgrade 

FEMA-350 Criteria for Existing Welded 

Appendix A: Detailed Procedures for Performance Evaluation Steel Moment-Frame Buildings 

large enough to allow P-D instabilities to develop. Prediction of the onset of P-D 

instability due to ground shaking is complex. It is possible that ground shaking 

will displace a structure to a point where static P-D instability would initiate, but 

will bring the structure back again before collapse can occur, due to a reversal in 

ground shaking direction. 

The basic effect of P-D instability is that a negative stiffness is induced in the 

structure. That is, P-D effects produce a condition in which increased 

displacement can occur at a reduced lateral force. A similar and equally 

dangerous effect can be produced by local hysteretic strength degradation of 

beam-column assemblies (FEMA-355C). Hysteretic strength degradation 

typically occurs after the onset of significant local buckling in the beam-column 

assemblies. It is important when performing Incremental Dynamic Analyses that 

these local strength degradation effects, which show up as a concave curvature in 

the hysteretic loops in laboratory data, are replicated by the analytical model. 

Nonlinear analysis software that is currently commercially available is not, in 

general, able to model this behavior. It is possible to account for these effects 

approximately by increasing the amount of dead load on the structure, so as to 

produce artificially the appropriate negative stiffness. 

A-18
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings References, Bibliography, and Acronyms 

REFERENCES, BIBLIOGRAPHY, AND ACRONYMS 

This section contains references, additional bibliography and acronyms that are generally 

common to the set of reports, FEMA-350, FEMA-351, FEMA-352, and FEMA-353. Following 

the regular references are three sections containing ASTM Standards published by the American 

Society for Testing and Materials, West Conshohocken, Pennsylvania and listed numerically, 

AWS Specifications published by the American Welding Society, Miami, Florida, and listed 

numerically, FEMA Reports published by the Federal Emergency Management Agency, 

Washington, DC, and listed by report number, and SAC Reports published by the SAC Joint 

Venture, Sacramento, California, and listed by report number. 

References and Additional Bibliography. 

AISC, 1985, Specification for Structural Joints using ASTM A325 or A490 Bolts, American 

Institute of Steel Construction, Chicago, Illinois. 

AISC, 1989, Manual of Steel Construction, ASD, Ninth Edition, American Institute of Steel 

Construction, Chicago, Illinois. 

AISC, 1993, 1997, Load and Resistance Factor Design Specifications for Structural Steel 

Buildings, American Institute of Steel Construction, Chicago, Illinois. 

AISC, 1994a, Proceedings of the AISC Special Task Committee on the Northridge Earthquake 

Meeting, American Institute of Steel Construction, Chicago, Illinois. 

AISC, 1994b, Northridge Steel Update 1, American Institute of Steel Construction, Chicago, 

Illinois. 

AISC, 1997, Seismic Provisions for Structural Steel Buildings, American Institute of Steel 

Construction, Chicago, Illinois. 

AISC, 1998a, Load and Resistance Factor Design Specifications for Structural Steel Buildings, 

American Institute of Steel Construction, Chicago, Illinois. 

AISC, 1998b, LRFD Manual of Steel Construction, 2nd Edition, American Institute of Steel 

Construction, Chicago, Illinois. 

AISC, 1999, Supplement No. 1 to the 1997 Seismic Provisions for Structural Steel Buildings, 

American Institute of Steel Construction, Chicago, Illinois. 

Allen, J., Partridge, J.E., Richard, R.M., and Radau, S., 1995, “Ductile Connection Designs for 

Welded Steel Moment Frames,” Proceedings, 64th Annual Convention, Structural Engineers 

Association of California, Sacramento, California. 

Anderson, J, Duan, J., Xiao, Y., and Maranian, P., 2000, Improvement of Welded Connections 

Using Fracture Tough Overlays, Report No. SAC/BD-00/20, SAC Joint Venture, 

Sacramento, California. 

ASCE, 1998, ASCE-7 maps, American Society of Civil Engineers, Reston, Virginia. 

ASTM citations: see the list of ASTM Standards on page R-4. 

R-1
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

References, Bibliography, and Acronyms Moment-Frame Buildings 

ATC, 1985, Earthquake Damage Evaluation Data for California, Report ATC-13 , Applied 

Technology Council, Redwood City, California. 

ATC, 1987, Evaluating the Seismic Resistance of Existing Buildings, Report ATC-14, Applied 

Technology Council, Redwood City, California. 

ATC, 1989, Procedures for Postearthquake Safety Evaluations of Buildings, Report ATC-20, 

Applied Technology Council, Redwood City, California. 

ATC, 1992, Guidelines for Cyclic Seismic Testing of Components of Steel Structures, Report 

ATC-24, Applied Technology Council, Redwood City, California. 

ATC, 1995, Addendum to the ATC-20 Postearthquake Building Safety Evaluation Procedures, 

Report ATC-20-2, Applied Technology Council, Redwood City, California. 

ATC, 1997a, Seismic Evaluation and Retrofit of Concrete Buildings, prepared by the Applied 

Technology Council (Report No. ATC-40), for the California Seismic Safety Commission 

(Report No. SSC 96-01), Sacramento, California. 

ATC, 1997b, NEHRP Guidelines for the Seismic Rehabilitation of Buildings, Report No. FEMA273, prepared by the Applied Technology Council for the Building Seismic Safety Council, 

published by the Federal Emergency Management Agency, Washington, DC. 

ATC, 1997c, Commentary to NEHRP Guidelines for the Seismic Rehabilitation of Buildings, 

Report No. FEMA-274, prepared by the Applied Technology Council for the Building 

Seismic Safety Council, published by the Federal Emergency Management Agency, 

Washington, DC. 

Avent, R., 1992, “Designing Heat-Straightening Repairs,” National Steel Construction 

Conference Proceedings, Las Vegas, Nevada. 

AWS citations: see the list of AWS reports, specifications and codes on page R-5. 

Barsom, J.M., 1996, “Steel Properties — Effects of Constraint, Temperature, and Rate of 

Loading,” Proceedings of the 2nd US Seminar, Seismic Design, Evaluation and Retrofit of 

Steel Bridges, San Francisco, Report No. UCB/CEE STEEL-96/09, Dept. of Civil and 

Environmental Engineering, UC Berkeley, pp.115-143. 

Boore, D.M., and Joyner, W.B., 1994, Proceedings of Seminar on New Developments in 

Earthquake Ground-Motion Estimation and Implications for Engineering Design Practice, 

Report ATC-35-1, Applied Technology Council, Redwood City, California, pp 6-1 to 6-41. 

BSSC, 1992, NEHRP Handbook for the Seismic Evaluation of Existing Buildings, developed by 

the Building Seismic Safety Council for the Federal Emergency Management Agency, 

Report FEMA-178, Washington, D.C. 

BSSC, 1997a, 1997 NEHRP Recommended Provisions for Seismic Regulations for New 

Buildings and Other Structures, Part 1 – Provisions, prepared by the Building Seismic 

Safety Council for the Federal Emergency Management Agency (Report No. FEMA-302), 

Washington, DC. 

BSSC, 1997b, 1997 NEHRP Recommended Provisions for Seismic Regulations for New 

Buildings and Other Structures, Part 2 – Commentary, prepared by the Building Seismic 

R-2
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings References, Bibliography, and Acronyms 

Safety Council for the Federal Emergency Management Agency (Report No. FEMA-303), 

Washington, DC. 

Campbell, K.W., and Bozorgnia, Y., 1994, “Near-Source Attenuation of Peak Horizontal 

Acceleration from Worldwide Accelerograms Recorded from 1957 to 1993,” Fifth U.S. 

National Conference on Earthquake Engineering, Proceedings, Vol. III, pp 283-292, 

Earthquake Engineering Research Institute, Oakland, California. 

Chi, W.M., Deierlein, G., and Ingraffea, A., 1997, “Finite Element Fracture Mechanics 

Investigation of Welded Beam-Column Connections,” SAC Joint Venture, Report No. 

SAC/BD-97/05. 

FEMA citations: see the list of FEMA reports on page R-6. 

Goel, R.K., and Chopra, A.K., 1997, “Period Formulas for Moment-Resisting Frame Buildings,” 

Journal of Structural Engineering, Vol. 123, No. 11, pp. 1454-1461. 

Gross, J.L., Engelhardt, M.D., Uang, C.M., Kasai, K. and Iwankiw, N.R., 1999, Modification of 

Existing Welded Steel Moment Frame Connections for Seismic Resistance, AISC Design 

Guide Series 12, American Institute of Steel Construction, Chicago, Illinois. 

Grubbs, K., 1997, “The Effect of the Dogbone Connection on the Elastic Stiffness of Steel 

Moment Frames” Masters Thesis, Department of Civil Engineering, University of Texas at 

Austin. 

ICBO, 1988, 1991, and 1997, Uniform Building Code, indicated edition, International 

Conference of Building Officials, Whittier, California. 

ICC, 2000, International Building Code, International Code Council, Falls Church, Virginia. 

Kircher, C.A., Nassar, A.A., Kustu, O. and Holmes, W.T., 1997, “Development of Building 

Damage Functions for Earthquake Loss Estimation,” Earthquake Spectra, Vol. 13, No. 4, 

Earthquake Engineering Research Institute, Oakland, California, pp. 663-682. 

Kircher, C.A., Reitherman, R.K., Whitman, R.V., and Arnold, C., 1997, “Estimation of 

Earthquake Losses to Buildings,” Earthquake Spectra, Vol. 13, No. 4, Earthquake 

Engineering Research Institute, Oakland, California, pp. 703-720. 

Kircher, C.A., 1999, Procedures for Development of HAZUS-Compatible Building-Specific 

Damage and Loss Functions, National Institute of Building Sciences, Washington, D.C. 

Krawinkler, H., Gupta, A., Medina, R. and Luco, N., 2000, Loading Histories for Seismic 

Performance Testing of SMRF Components and Assemblies, Report No. SAC/BD-00/10, 

SAC Joint Venture, Sacramento, California. 

NIBS, 1997a, HAZUS Earthquake Loss Estimation Methodology, Users Manual, National 

Institute of Building Sciences, Washington, DC. 

NIBS, 1997b, HAZUS Earthquake Loss Estimation Methodology, Technical Manual, 3 Volumes. 

National Institute of Building Sciences, Washington, DC. 

RCSC, 1996, Load and Resistance Factor Design: Specification for Structural Joints Using 

ASTM A325 or A490 Bolts, Research Council on Structural Connections. 

R-3
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

References, Bibliography, and Acronyms Moment-Frame Buildings 

Richard, R., Partridge, J.E., Allen, J., and Radau, S., 1995, “Finite Element Analysis and Tests of 

Beam-to-Column Connections,” Modern Steel Construction, Vol. 35, No. 10, pp. 44-47, 

American Institute of Steel Construction, Chicago, Illinois. 

SAC citations: see the list of SAC Joint Venture reports on page R-7. 

Shonafelt, G.O., and Horn, W.B, 1984, Guidelines for Evaluation and Repair of Damaged Steel 

Bridge Members, NCHRP Report 271, prepared by the National Cooperative Highway 

Research Program, for the Transportation Research Board, Washington, DC. 

Wald, D.J., Quitoriano, T.H., Kanamori, H. and Scrivner, C.W., 1998, “Trinet Shakemaps – 

Rapid Generation of Peak Ground Motion and Intensity Maps for Earthquakes in Southern 

California”, SMIP98 Proceedings, California Division of Mines and Geology, Sacramento, 

California. 

Whitman, R., Anagnos, T., Kircher, C., Lagorio, H.J., Lawson, R.S., and Schneider, P., 1997, 

“Development of a National Earthquake Loss-Estimation Methodology,” Earthquake 

Spectra, Vol. 13, No. 4, Earthquake Engineering Research Institute, Oakland, California, 

pp. 643-661. 

Youssef, N.F.G, Bonowitz, D., and Gross, J.L., 1995, A Survey of Steel Moment-Resisting Frame 

Buildings Affected by the 1994 Northridge Earthquake, Report No. NISTR 56254, National 

Institute for Science and Technology, Gaithersburg, Maryland. 

ASTM Standards. 

ASTM Standards are published by the American Society for Testing and Materials, West 

Conshohocken, Pennsylvania, and are listed alphanumerically. 

ASTM, 1997, Standard Test Methods and Definitions for Mechanical Testing of Steel Products 

A6, Supplementary Requirement S5 

A36, Specification for Carbon Structural Steel 

A325, Specification for Structural Bolts, Steel, Heat-Treated, 120/105 ksi Minimum Tensile 

Strength 

A435, Straight Beam Ultrasonic Examination of Steel Plates 

A490, Specification for Heat-Treated Steel Structural Bolts, 150 ksi Minimum Tensile Strength 

A563, Specification for Carbon and Alloy Steel Nuts 

A572, Specification for High-Strength Low-Alloy Columbium-Vanadium Structural Steel 

A898, Straight Beam Ultrasonic Examination of Rolled Steel Structural Shapes 

A913, Specification for High-Strength Low-Alloy Steel Shapes of Structural Quality, Produced 

by Quenching and Self-Tempering Process 

A992, Standard Specification for Steel for Structural Shapes for Use in Building Framing 

E329, Standard Specification for Agencies Engaged in the Testing and/or Inspection of Material 

Used in Construction 

R-4
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings References, Bibliography, and Acronyms 

E543, Standard Practice for Agencies Performing Nondestructive Testing 

E548, Standard Guide for General Criteria Used for Evaluating Laboratory Competence 

E994, Standard Guide for Laboratory Accreditation Systems 

E1212, Standard Practice for Establishment and Maintenance of Quality Control Systems for 

Nondestructive Testing Agencies 

E1359, Standard Guide for Surveying Nondestructive Testing Agencies 

F436, Specification for Hardened Steel Washers 

F959, Specification for Compressible-Washer-Type Direct Tension Indicators for Use with 

Structural Fasteners 

F1554, Specification for Anchor Bolts, Steel, 36, 55, and 105 ksi Yield Strength 

F1852, Specification for “Twist-Off” Type Tension Control Structural Bolt/Nut/Washer 

Assemblies, Steel, Heat Treated, 120/105 ksi Minimum Tensile Strength 

AWS Reports, Specifications, and Codes. 

AWS reports are published by the American Welding Society, Miami, Florida, and are listed 

alphanumerically. 

AWS A2.4, Standard Symbols for Welding, Brazing, and Nondestructive Testing 

AWS A4.3, Standard Methods for Determination of the Diffusible Hydrogen Content of 

Martensitic, Bainitic, and Ferritic Steel Weld Metal Produced by Arc Welding 

ANSI/AWS A5.1-91, Specification for Carbon Steel Electrodes for Shielded Metal Arc Welding 

ANSI/AWS A5.18-93, Specification for Carbon Steel Electrodes and Rods for Gas Shielded Arc 

Welding 

ANSI/AWS A5.20-95, Specification for Carbon Steel Electrodes for Flux-Cored Arc Welding 

AWS, 1995, Presidential Task Group Report 

ANSI/AWS A5.5-96, Specification for Low-Alloy Steel Electrodes for Shielded Metal Arc 

Welding 

ANSI/AWS A5.28-96, Specification for Low-Alloy Steel Electrodes and Rods for Gas Shielded 

Arc Welding 

ANSI/AWS A5.23/A5.23M-97, Specification for Low-Alloy Steel Electrodes and Fluxes for 

Submerged Arc Welding 

ANSI/AWS A5.25/A5.25M-97, Specification for Carbon and Low-Alloy Steel Electrodes and 

Fluxes for Electroslag Welding 

ANSI/AWS A5.26/A5.26M-97, Specification for Carbon and Low-Alloy Steel Electrodes for 

Electrogas Welding 

ANSI/AWS A5.32/A5.32M-97, Specification for Welding Shielding Gases 

R-5
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

References, Bibliography, and Acronyms Moment-Frame Buildings 

ANSI/AWS A5.17/A5.17M-97, Specification for Carbon Steel Electrodes and Fluxes for 

Submerged Arc Welding 

ANSI/AWS A5.29-98, Specification for Low-Alloy Steel Electrodes for Flux-Cored Arc Welding 

AWS D1.1-1998, 2000, Structural Welding Code – Steel 

AWS D1.3, Structural Welding Code 

AWS D1.4, Structural Welding Code 

AWS QC1, Standard for AWS Certification of Welding Inspectors 

FEMA Reports. 

FEMA reports are listed by report number. 

FEMA-178, 1992, NEHRP Handbook for the Seismic Evaluation of Existing Buildings, 

developed by the Building Seismic Safety Council for the Federal Emergency Management 

Agency, Washington, DC. 

FEMA-267, 1995, Interim Guidelines, Inspection, Evaluation, Repair, Upgrade and Design of 

Welded Moment Resisting Steel Structures, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Washington, DC. 

FEMA-267A, 1996, Interim Guidelines Advisory No. 1, prepared by the SAC Joint Venture for 

the Federal Emergency Management Agency, Washington, DC. 

FEMA-267B, 1999, Interim Guidelines Advisory No. 2, prepared by the SAC Joint Venture for 

the Federal Emergency Management Agency, Washington, DC. 

FEMA-273, 1997, NEHRP Guidelines for the Seismic Rehabilitation of Buildings, prepared by 

the Applied Technology Council for the Building Seismic Safety Council, published by the 

Federal Emergency Management Agency, Washington, DC. 

FEMA-274, 1997, NEHRP Commentary on the Guidelines for the Seismic Rehabilitation of 

Buildings, prepared by the Applied Technology Council for the Building Seismic Safety 

Council, published by the Federal Emergency Management Agency, Washington, DC. 

FEMA-302, 1997, NEHRP Recommended Provisions for Seismic Regulations for New Buildings 

and Other Structures, Part 1 – Provisions, prepared by the Building Seismic Safety Council 

for the Federal Emergency Management Agency, Washington, DC. 

FEMA-303, 1997, NEHRP Recommended Provisions for Seismic Regulations for New Buildings 

and Other Structures, Part 2 – Commentary, prepared by the Building Seismic Safety 

Council for the Federal Emergency Management Agency, Washington, DC. 

FEMA-310, 1998, Handbook for the Seismic Evaluation of Buildings – A Prestandard, prepared 

by the American Society of Civil Engineers for the Federal Emergency Management 

Agency, Washington, DC. 

FEMA-350, 2000, Recommended Seismic Design Criteria for New Steel Moment-Frame 

Buildings, prepared by the SAC Joint Venture for the Federal Emergency Management 

Agency, Washington, DC. 

R-6
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings References, Bibliography, and Acronyms 

FEMA-351, 2000, Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded 

Steel Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

FEMA-352, 2000, Recommended Postearthquake Evaluation and Repair Criteria for Welded 

Steel Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

FEMA-353, 2000, Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications, prepared by the SAC Joint Venture 

for the Federal Emergency Management Agency, Washington, DC. 

FEMA-354, 2000, A Policy Guide to Steel Moment-Frame Construction, prepared by the SAC 

Joint Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355A, 2000, State of the Art Report on Base Metals and Fracture, prepared by the SAC 

Joint Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355B, 2000, State of the Art Report on Welding and Inspection, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355C, 2000, State of the Art Report on Systems Performance of Steel Moment Frames 

Subject to Earthquake Ground Shaking, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

FEMA-355D, 2000, State of the Art Report on Connection Performance, prepared by the SAC 

Joint Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355E, 2000, State of the Art Report on Past Performance of Steel Moment-Frame 

Buildings in Earthquakes, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

FEMA-355F, 2000, State of the Art Report on Performance Prediction and Evaluation of Steel 

Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC Joint Venture Reports. 

SAC Joint Venture reports are listed by report number, except for SAC 2000a through 2000k; 

those entries that do not include a FEMA report number are published by the SAC Joint 

Venture. 

SAC 94-01, 1994, Proceedings of the Invitational Workshop on Steel Seismic Issues, Los 

Angeles, September 1994, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC 94-01, 1994b, Proceedings of the International Workshop on Steel Moment Frames, 

Sacramento, December, 1994, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC 95-01, 1995, Steel Moment Frame Connection Advisory No. 3, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Washington, DC. 

R-7
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

References, Bibliography, and Acronyms Moment-Frame Buildings 

SAC 95-02, 1995, Interim Guidelines: Evaluation, Repair, Modification and Design of Welded 

Steel Moment Frame Structures, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Report No. FEMA-267, Washington, DC. 

SAC 95-03, 1995, Characterization of Ground Motions During the Northridge Earthquake of 

January 17, 1994, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC 95-04, 1995, Analytical and Field Investigations of Buildings Affected by the Northridge 

Earthquake of January 17, 1994, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

SAC 95-05, 1995, Parametric Analytic Investigations of Ground Motion and Structural 

Response, Northridge Earthquake of January 17, 1994, prepared by the SAC Joint Venture 

for the Federal Emergency Management Agency, Washington, DC. 

SAC 95-06, 1995, Technical Report: Surveys and Assessment of Damage to Buildings Affected 

by the Northridge Earthquake of January 17, 1994, prepared by the SAC Joint Venture for 

the Federal Emergency Management Agency, Washington, DC. 

SAC 95-07, 1995, Technical Report: Case Studies of Steel Moment-Frame Building 

Performance in the Northridge Earthquake of January 17, 1994, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Washington, DC. 

SAC 95-08, 1995, Experimental Investigations of Materials, Weldments and Nondestructive 

Examination Techniques, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC 95-09, 1995, Background Reports: Metallurgy, Fracture Mechanics, Welding, Moment 

Connections and Frame Systems Behavior, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Report No. FEMA-288, Washington, DC. 

SAC 96-01, 1996, Experimental Investigations of Beam-Column Subassemblages, Part 1 and 2, 

prepared by the SAC Joint Venture for the Federal Emergency Management Agency, 

Washington, DC. 

SAC 96-02, 1996, Connection Test Summaries, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Report No. FEMA-289, Washington, DC. 

SAC 96-03, 1997, Interim Guidelines Advisory No. 1 Supplement to FEMA-267 Interim 

Guidelines, prepared by the SAC Joint Venture for the Federal Emergency Management 

Agency, Report No. FEMA-267A, Washington, DC. 

SAC 98-PG, Update on the Seismic Safety of Steel Buildings – A Guide for Policy Makers, 

prepared by the SAC Joint Venture for the Federal Emergency Management Agency, 

Washington, DC. 

SAC 99-01, 1999, Interim Guidelines Advisory No. 2 Supplement to FEMA-267 Interim 

Guidelines, prepared by the SAC Joint Venture, for the Federal Emergency Management 

Agency, Report No. FEMA-267B, Washington, DC. 

R-8
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings References, Bibliography, and Acronyms 

SAC, 2000a, Recommended Seismic Design Criteria for New Steel Moment-Frame Buildings, 

prepared by the SAC Joint Venture for the Federal Emergency Management Agency, Report 

No. FEMA-350, Washington, D.C. 

SAC, 2000b, Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded Steel 

Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Report No. FEMA-351, Washington, D.C. 

SAC, 2000c, Recommended Postearthquake Evaluation and Repair Criteria for Welded Steel 

Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Report No. FEMA-352, Washington, D.C. 

SAC, 2000d, Recommended Specifications and Quality Assurance Guidelines for Steel MomentFrame Construction for Seismic Applications, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Report No. FEMA-353, Washington, D.C. 

SAC, 2000e, A Policy Guide to Steel Moment-Frame Construction, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-354, 

Washington, D.C. 

SAC, 2000f, State of the Art Report on Base Metals and Fracture, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355A, 

Washington, D.C. 

SAC, 2000g, State of the Art Report on Welding and Inspection, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355B, 

Washington, D.C. 

SAC, 2000h, State of the Art Report on Systems Performance, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355C, 

Washington, D.C. 

SAC, 2000i, State of the Art Report on Connection Performance, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355D, 

Washington, D.C. 

SAC, 2000j, State of the Art Report on Past Performance of Steel Moment-Frame Buildings in 

Earthquakes, prepared by the SAC Joint Venture for the Federal Emergency Management 

Agency, Report No. FEMA-355E, Washington, D.C. 

SAC, 2000k, State of the Art Report on Performance Prediction and Evaluation, prepared by the 

SAC Joint Venture for the Federal Emergency Management Agency, Report No. FEMA355F, Washington, D.C. 

SAC/BD-96/01, Selected Results from the SAC Phase 1 Beam-Column Connection Pre-Test 

Analyses, submissions from B. Maison, K. Kasai, and R. Dexter; and A. Ingraffea and G. 

Deierlein. 

SAC/BD-96/02, Summary Report on SAC Phase 1 - Task 7 Experimental Studies, by C. Roeder 

(a revised version of this document is published in Report No. SAC 96-01; the original is no 

longer available). 

SAC/BD-96/03, Selected Documents from the U.S.-Japan Workshop on Steel Fracture Issues. 

R-9
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

References, Bibliography, and Acronyms Moment-Frame Buildings 

SAC/BD-96/04, Survey of Computer Programs for the Nonlinear Analysis of Steel Moment 

Frame Structures. 

SAC/BD-97/01, Through-Thickness Properties of Structural Steels, by J. Barsom and S. 

Korvink. 

SAC/BD-97/02, Protocol for Fabrication, Inspection, Testing, and Documentation of BeamColumn Connection Tests and Other Experimental Specimens, by P. Clark, K. Frank, H. 

Krawinkler, and R. Shaw. 

SAC/BD-97/03, Proposed Statistical and Reliability Framework for Comparing and Evaluating 

Predictive Models for Evaluation and Design, by Y.-K. Wen. 

SAC/BD-97/04, Development of Ground Motion Time Histories for Phase 2 of the FEMA/SAC 

Steel Project, by. P. Somerville, N. Smith, S. Punyamurthula, and J. Sun. 

SAC/BD-97/05, Finite Element Fracture Mechanics Investigation of Welded Beam-Column 

Connections, by W.-M. Chi, G. Deierlein, and A. Ingraffea. 

SAC/BD-98/01, Strength and Ductility of FR Welded-Bolted Connections, by S. El-Tawil, T. 

Mikesell, E. Vidarsson, and S. K. Kunnath. 

SAC/BD-98/02, Effects of Strain Hardening and Strain Aging on the K-Region of Structural 

Shapes, by J. Barsom and S. Korvink 

SAC/BD-98/03, Implementation Issues for Improved Seismic Design Criteria: Report on the 

Social, Economic, Policy and Political Issues Workshop by L.T. Tobin. 

SAC/BD-99/01, Parametric Study on the Effect of Ground Motion Intensity and Dynamic 

Characteristics on Seismic Demands in Steel Moment Resisting Frames by G. A. MacRae 

SAC/BD-99/01A, Appendix to: Parametric Study on the Effect of Ground Motion Intensity and 

Dynamic Characteristics on Seismic Demands in Steel Moment Resisting Frames by G. A. 

MacRae 

SAC/BD-99/02, Through-Thickness Strength and Ductility of Column Flange in Moment 

Connections by R. Dexter and M. Melendrez. 

SAC/BD-99/03, The Effects of Connection Fractures on Steel Moment Resisting Frame Seismic 

Demands and Safety by C. A. Cornell and N. Luco 

SAC/BD-99/04, Effects of Strength/Toughness Mismatch on Structural and Fracture Behaviors 

in Weldments by P. Dong, T. Kilinski, J. Zhang and F.W. Brust 

SAC/BD-99/05, Assessment of the Reliability of Available NDE Methods for Welded Joint and 

the Development of Improved UT Procedures by G. Gruber and G. Light 

SAC/BD-99/06, Prediction of Seismic Demands for SMRFs with Ductile Connections and 

Elements by A. Gupta and H. Krawinkler 

SAC/BD-99/07, Characterization of the Material Properties of Rolled Sections by T. K. Jaquess 

and K. Frank 

SAC/BD-99/08, Study of the Material Properties of the Web-Flange Intersection of Rolled 

Shapes by K. R. Miller and K. Frank 

R-10
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings References, Bibliography, and Acronyms 

SAC/BD-99/09, Investigation of Damage to WSMF Earthquakes other than Northridge by M. 

Phipps 

SAC/BD-99/10, Clarifying the Extent of Northridge Induced Weld Fracturing and Examining 

the Related Issue of UT Reliability by T. Paret 

SAC/BD-99/11, The Impact of Earthquakes on Welded steel Moment Frame Buildings: 

Experience in Past Earthquakes by P. Weinburg and J. Goltz 

SAC/BD-99/12, Assessment of the Benefits of Implementing the New Seismic Design Criteria 

and Inspection Procedures by H. A. Seligson and R. Eguchi 

SAC/BD-99/13, Earthquake Loss Estimation for WSMF Buildings, by C. A. Kircher 

SAC/BD-99/14, Simplified Loss Estimation for Pre-Northridge WSMF Buildings, by B. F. 

Maison and D. Bonowitz 

SAC/BD-99/15, Integrative Analytical Investigations on the Fracture Behavior of Welded 

Moment Resisting Connections, by G. G. Deierlein and W.-M. Chi 

SAC/BD-99/16, Seismic Performance of 3 and 9 Story Partially Restrained Moment Frame 

Buildings, by B. F. Maison and K. Kasai 

SAC/BD-99/17, Effects of Partially-Restrained Connection Stiffness and Strength on Frame 

Seismic Performance, by K. Kasai, B. F. Maison, and A. Mayangarum 

SAC/BD-99/18, Effects of Hysteretic Deterioration Characteristics on Seismic Response of 

Moment Resisting Steel Structures, by F. Naeim, K. Skliros, A. M. Reinhorn and M.V. 

Sivaselvan 

SAC/BD-99/19, Cyclic Instability of Steel Moment Connections with Reduced Beam Section, by 

C.-M. Uang and C.-C. Fan 

SAC/BD-99/20, Local and Lateral-Torsion Buckling of Wide Flange Beams, by L. 

Kwasniewski, B. Stojadinovic, and S. C. Goel 

SAC/BD-99/21, Elastic Models for Predicting Building Performance, by X. Duan and J. C. 

Anderson 

SAC/BD-99/22, Reliability-Based Seismic Performance Evaluation of Steel Frame Buildings 

Using Nonlinear Static Analysis Methods, by G. C. Hart and M. J. Skokan 

SAC/BD-99/23, Failure Analysis of Welded Beam to Column Connections, by J. M. Barsom 

SAC/BD-99/24, Weld Acceptance Criteria for Seismically-Loaded Welded Connections, by W. 

Mohr 

SAC/BD-00/01, Parametric Tests on Unreinforced Connections, by K.-H. Lee, B. Stojadinovic, 

S. C. Goel, A. G. Margarian, J. Choi, A. Wongkaew, B. P. Reyher, and D.-Y, Lee 

SAC/BD-00/02, Parametric Tests on the Free Flange Connections, by J. Choi, B. Stojadinovic, 

and S. C. Goel 

SAC/BD-00/03, Cyclic Tests on Simple Connections Including Effects of the Slab, by J. Liu and 

A. Astaneh-Asl 

R-11
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

References, Bibliography, and Acronyms Moment-Frame Buildings 

SAC/BD-00/04, Tests on Bolted Connections, by J. Swanson, R. Leon and J. Smallridge 

SAC/BD-00/05, Bolted Flange Plate Connections, by S. P. Schneider and I. Teeraparbwong 

SAC/BD-00/06, Round Robin Testing of Ultrasonic Testing Technicians, by R. E. Shaw, Jr. 

SAC/BD-00/07, Dynamic Tension Tests of Simulated Welded Beam Flange Connections, by J. 

M. Ricles, C. Mao, E. J. Kaufmann, L.-W. Lu, and J. W. Fisher 

SAC/BD-00/08, Design of Steel Moment Frame Model Buildings in Los Angeles, Seattle and 

Boston, by P. Clark 

SAC/BD-00/09, Benchmarking of Analysis Programs for SMRF System Performance Studies, by 

A. G. and H. Krawinkler 

SAC/BD-00/10, Loading Histories for Seismic Performance Testing of SMRF Components and 

Assemblies, by H. Krawinkler, A. Gupta, R. Medina and N. Luco 

SAC/BD-00/11, Development of Improved Post-Earthquake Inspection Procedures for Steel 

Moment Frame Buildings, by P. Clark 

SAC/BD-00/12, Evaluation of the Effect of Welding Procedure on the Mechanical Properties of 

FCAW-S and SMAW Weld Metal Used in the Construction of Seismic Moment Frames, by 

M. Q. Johnson 

SAC/BD-00/13, Preliminary Evaluation of Heat Affected Zone Toughness in Structural Shapes 

Used in the Construction of Seismic Moment Frames, by M. Q. Johnson 

SAC/BD-00/14, Evaluation of Mechanical Properties in Full-Scale Connections and 

Recommended Minimum Weld Toughness for Moment Resisting Frames, by M. Q. Johnson, 

W. Mohr, and J. Barsom 

SAC/BD-00/15, Simplified Design Models for Predicting the Seismic Performance of Steel 

Moment Frame Connections, by C. Roeder, R.G. Coons, and M. Hoit 

SAC/BD-00/16, SAC Phase 2 Test Plan, by C. Roeder 

SAC/ BD-00/17, Behavior and Design of Radius-Cut, Reduced Beam Section Connections, by 

M. Engelhardt, G. Fry, S. Johns, M. Venti, and S. Holliday 

SAC/BD-00/18, Test of a Free Flange Connection with a Composite Floor Slab, by M. Venti 

and M. Engelhardt 

SAC/BD-00/19, Cyclic Testing of a Free Flange Moment Connection by C. Gilton, B. Chi, and 

C. M. Uang 

SAC/BD-00/20, Improvement of Welded Connections Using Fracture Tough Overlays, by James 

Anderson, J. Duan, P. Maranian, and Y. Xiao 

SAC/BD-00/21, Cyclic Testing of Bolted Moment End-Plate Connections, by T. Murray and E. 

Sumner 

SAC/BD-00/22, Cyclic Response of RBS Moment Connections: Loading Sequence and Lateral 

Bracing Effects, by Q.S. Yu, C. Gilton, and C. M. Uang 

R-12
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings References, Bibliography, and Acronyms 

SAC/BD-00/23, Cyclic Response of RBS Moment Connections: Weak Axis Configuration and 

Deep Column Effects, by C. Gilton, B. Chi, and C. M. Uang 

SAC/BD-00/24, Development and Evaluation of Improved Details for Ductile Welded 

Unreinforced Flange Connections, by J.M. Ricles, C. Mao, L.W. Lu, and J. Fisher 

SAC/BD-00/25, Performance Prediction and Evaluation of Steel Special Moment Frames for 

Seismic Loads, by K. Lee and D. A. Foutch 

SAC/BD-00/26, Performance Prediction and Evaluation of Low Ductility Steel Moment Frames 

for Seismic Loads, by S. Yun and D. A. Foutch 

SAC/BD-00/27, Steel Moment Resisting Connections Reinforced with Cover and Flange Plates, 

by T. Kim, A.S. Whittaker, V.V. Bertero, and A.S.J. Gilani 

SAC/BD-00/28, Failure of a Column K-Area Fracture, by J.M. Barsom and J.V. Pellegrino 

SAC/BD-00/29, Inspection Technology Workshop, by R. E. Shaw, Jr. 

Acronyms. 

A, acceleration response

ACAG, air carbon arc gouging

ACIL, American Council of Independent

Laboratories 

AISC, American Institute for Steel 

Construction 

ANSI, American National Standards Institute 

API, American Petroleum Institute 

ASNT, American Society for Nondestructive 

Testing 

ASTM, American Society for Testing and 

Materials 

ATC, Applied Technology Council 

A2LA, American Association for Laboratory 

Accreditation 

AWS, American Welding Society 

BB, Bolted Bracket (connection) 

BFP, Bolted Flange Plates (connection) 

BOCA, Building Officials and Code 

Administrators 

BSEP, Bolted Stiffened End Plate 

(connection) 

BUEP, Bolted Unstiffened End Plate 

(connection) 

CAC-A, air carbon arc cutting 

CAWI, Certified Associate Welding Inspector 

CJP, complete joint penetration (weld) 

CP, Collapse Prevention (performance level) 

CUREe, California Universities for Research 

in Earthquake Engineering 

CVN, Charpy V-notch 

CWI, Certified Welding Inspector 

D, displacement response 

DST, Double Split Tee (connection) 

DTI, Direct Tension Indicator 

EGW, electrogas welding 

ELF, equivalent lateral force 

ESW, electroslag welding 

FCAW-S, flux-cored arc welding – selfshielded 

FCAW-G, flux-cored arc welding – gasshielded 

FEMA, Federal Emergency Management 

Agency 

FF, Free Flange (connection) 

FR, fully restrained (connection) 

GMAW, gas metal arc welding 

GTAW, gas tungsten arc welding 

HAZ, heat-affected zone 

IBC, International Building Code 

ICBO, International Conference of Building 

Officials 

ICC, International Code Council 

IMF, Intermediate Moment Frame 

IO, Immediate Occupancy (performance 

level) 

R-13
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

References, Bibliography, and Acronyms Moment-Frame Buildings 

ISO, International Standardization 

Organization 

IWURF, Improved Welded Unreinforced 

Flange (connection) 

L, longitudinal 

LDP, Linear Dynamic Procedure 

LRFD, load and resistance-factor design 

LS, Life Safety (performance level) 

LSP, Linear Static Procedure 

MCE, Maximum Considered Earthquake 

MMI, Modified Mercalli Intensity 

MRS, modal response spectrum 

MRSF, steel moment frame 

MT, magnetic particle testing 

NBC, National Building Code 

NDE, nondestructive examination 

NDP, Nonlinear Dynamic Procedure 

NDT, nondestructive testing 

NEHRP, National Earthquake Hazard 

Reduction Program 

NES, National Evaluation Services 

NSP, Nonlinear Static Procedure 

NVLAP, National Volunteer Laboratory 

Accreditation Program 

OMF, Ordinary Moment Frame 

PGA, peak ground acceleration 

PJP, partial joint penetration (weld) 

PIDR, pseudo interstory drift ratio 

PQR, Performance Qualification Record 

PR, partially restrained (connection) 

PT, liquid dye penetrant testing 

PWHT, postweld heat treatment 

PZ, panel zone 

QA, quality assurance 

QC, quality control 

QCP, Quality Control Plan, Quality 

Certification Program 

RBS, Reduced Beam Section (connection) 

RCSC, Research Council for Structural 

Connections 

RT, radiographic testing 

SAC, the SAC Joint Venture; a partnership of 

the Structural Engineers Association of 

California, the Applied Technology 

Council, and California Universities for 

Research in Earthquake Engineering 

SAW, submerged arc welding

SBC, Standard Building Code

SBCCI, Southern Building Code Congress

International 

SCWI, Senior Certified Welding Inspector 

SEAOC, Structural Engineers Association of 

California 

SFRS, seismic-force-resisting system 

SMAW, shielded metal arc welding 

SMF, Special Moment Frame 

SP, Side Plate (connection) 

SUG, Seismic Use Group 

SW, Slotted Web (connection) 

T, transverse 

TIGW, tungsten inert gas welding 

UBC, Uniform Building Code 

UT, ultrasonic testing 

VI, visual inspection 

WBH, Welded Bottom Haunch (connection) 

WCPF, Welded Cover Plate Flange 

(connection) 

WFP, Welded Flange Plate (connection) 

WPQR, Welding Performance Qualification 

Record 

WPS, Welding Procedure Specification 

WSMF, welded steel moment frame 

WT, Welded Top Haunch (connection) 

WTBH, Welded Top and Bottom Haunch 

(connection) 

WUF-B, Welded Unreinforced Flanges – 

Bolted Web (connection) 

WUF-W, Welded Unreinforced Flanges – 

Welded Web (connection) 

R-14
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings SAC Project Participants 

SAC Phase II Project Participants 

FEMA Project Officer FEMA Technical Advisor 

Michael Mahoney Robert D. Hanson

Federal Emergency Management Agency Federal Emergency Management Agency

500 C St. SW, Room 404 DFO Room 353

Washington, DC 20472 P.O. Box 6020

Pasadena, CA 91102-6020 

Joint Venture Management Committee (JVMC) 

William T. Holmes, Chair

Rutherford and Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

Edwin T. Huston

Smith & Huston, Inc.

8618 Roosevelt Way NE

Seattle, WA 98115

Robert Reitherman

California Universities for Research in

Earthquake Engineering 

1301 South 46th St. 

Richmond, CA 94804 

Christopher Rojahn

Applied Technology Council

555 Twin Dolphin Dr., Suite 550

Redwood City, CA 94065

Arthur E. Ross

Cole/Yee/Shubert & Associates

2500 Venture Oaks Way, Suite 100

Sacramento, CA 95833

Robin Shepherd

Earthquake Damage Analysis Corporation

40585 Lakeview Drive, Suite 1B

P.O. Box 1967

Big Bear Lake, CA 92315

Project Management Committee (PMC) 

Stephen A. Mahin, Project Manager

Pacific Earthquake Engr. Research Center

University of California

Berkeley, CA 94720

Ronald O. Hamburger, Project Director for

Project Development

EQE International

1111 Broadway, 10th Floor

Oakland, CA 94607-5500

James O. Malley, Project Director for

Topical Investigations

Degenkolb Engineers

225 Bush St., Suite 1000

San Francisco, CA 94104-1737

William T. Holmes, JVMC

Rutherford and Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

Christopher Rojahn, JVMC

Applied Technology Council

555 Twin Dolphin Dr., Suite 550

Redwood City, CA 94065

Robin Shepherd, JVMC

Earthquake Damage Analysis Corporation

40585 Lakeview Drive, Suite 1B

P.O. Box 1967

Big Bear Lake, CA 92315

S-1
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

SAC Project Participants Moment-Frame Buildings 

Peter W. Clark, Technical Assistant to PMC

SAC Steel Project Technical Office

1301 South 46th St.

Richmond, CA 94804

Project Administration 

Allen Paul Goldstein, Project Administrator Lori Campbell, Assistant to the Project 

A.P. Goldstein Associates

1621B 13th Street

Sacramento, CA 95814

Lee Adler

Structural Engineers Association of

California 

1730 I Street, Ste. 240 

Sacramento, CA 95814 

Administrator 

1621 B 13th Street 

Sacramento, CA 95628 

Project Oversight Committee (POC) 

William J. Hall, Chair

3105 Valley Brook Dr.

Champaign, IL 61821

Shirin Ader

International Conference of Building

Officials 

5360 Workman Mill Rd. 

Whittier, CA 90601-2298 

John M. Barsom

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

Roger Ferch

Herrick Corporation

7021 Koll Center Parkway

P.O Box 9125

Pleasanton, CA 94566-9125

Theodore V. Galambos

University of Minnesota

122 CE Building, 500 Pillsbury Dr. SE

Minnneapolis, MN 55455

John L. Gross

National Institute of Stds. & Technology

Building and Fire Research Lab,

Building 226, Room B158

Gaithersburg, MD 20899

James R. Harris

J.R. Harris and Co.

1580 Lincoln St., Suite 550

Denver, CO 80203-1509

Richard Holguin

520 Kathryn Ct.

Nipomo, CA 93444

Nestor Iwankiw

American Institute of Steel Construction

One East Wacker Dr., Suite 3100

Chicago, IL 60601-2001

Roy Johnston

Brandow & Johnston Associates

1600 West 3rd St.

Los Angeles, CA 90017

S-2
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings SAC Project Participants 

Leonard Joseph

Thornton-Tomassetti Engineers

641 6th Ave., 7th Floor

New York, NY 10011

Duane K. Miller

The Lincoln Electric Company

22801 St. Clair Ave.

Cleveland, OH 44117-1194

John Theiss

EQE/Theiss Engineers

1848 Lackland Hills Parkway

St. Louis, MO 63146-3572

John H. Wiggins

J.H. Wiggins Company

1650 South Pacific Coast Hwy, Suite 311

Redondo Beach, CA 90277

Team Leaders for Topical Investigations 

Douglas A. Foutch

University of Illinois

MC-250, 205 N. Mathews Ave.

3129 Newmark Civil Engineering Lab

Urbana, IL 61801

Karl H. Frank

University of Texas at Austin

10100 Bornet Rd.

Ferguson Lab, P.R.C. #177

Austin, TX 78758

Matthew Johnson

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

Helmut Krawinkler

Department of Civil Engineering

Stanford University

Stanford, CA 94305

Charles W. Roeder

University of Washington

233-B More Hall FX-10

Dept. of Building and Safety

Seattle, WA 98195-2700

L. Thomas Tobin

Tobin and Associates

134 California Ave.

Mill Valley, CA 94941

Lead Guideline Writers 

John D. Hooper

Skilling Ward Magnusson Barkshire, Inc.

1301 Fifth Avenue, Suite 3200

Seattle, WA 98101-2699

Lawrence D. Reaveley

University of Utah

Civil Engineering Dept.

3220 Merrill Engineering Building

Salt Lake City, UT 84112

Thomas A. Sabol

Englekirk & Sabol Consulting Engineers

P.O. Box 77-D

Los Angeles, CA 90007

C. Mark Saunders

Rutherford & Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

Robert E. Shaw

Steel Structures Technology Center, Inc.

42400 W Nine Mile Road

Novi, MI 48375-4132

Raymond H. R. Tide

Wiss, Janney, Elstner Associates, Inc.

330 Pfingsten Road

Northbrook, IL 60062-2095

S-3
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

SAC Project Participants Moment-Frame Buildings 

C. Allin Cornell, Associate Guideline Writer

Stanford University

Terman Engineering Center

Stanford, CA 94305-4020

Technical Advisory Panel (TAP) for Materials and Fracture 

John M. Barsom, POC

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

Serge Bouchard*

TradeARBED

825 Third Avenue, 35th Floor

New York, NY 10022

Michael F. Engestrom*

Nucor-Yamato Steel

P.O. Box 678

Frederick, MD 21705-0678

Karl H. Frank, Team Leader

University of Texas at Austin

10100 Bornet Rd.

Ferguson Lab, P.R.C. #177

Austin, TX 78758

Nestor Iwankiw*

American Institute of Steel Construction

One East Wacker Dr., Suite 3100

Chicago, IL 60601-2001

Dean C. Krouse*

705 Pine Top Drive

Bethelem, PA 18017

Frederick V. Lawrence

University of Illinois at Urbana-Champaign

205 N. Mathews Ave.

Room 2129 Newmark Lab

Urbana, IL 61801

Robert F. Preece

Preece, Goudie & Associates

100 Bush St., Suite 410

San Francisco, CA 94104

Raymond H. R. Tide, Guideline Writer

Wiss, Janney, Elstner Associates, Inc.

330 Pfingsten Road

Northbrook, IL 60062-2095

TAP for Welding and Inspection 

John M. Barsom

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

John W. Fisher

Lehigh University

117 ATLSS Drive

Bethlehem, PA 18015-4729

J. Ernesto Indacochea

University of Illinois at Chicago

Civil and Materials Engineering (mc 246)

842 West Taylor Street

Chicago, IL 60607

Matthew Johnson, Team Leader

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

S-4
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings SAC Project Participants 

David Long

PDM Strocal, Inc.

2324 Navy Drive

Stockton, CA 95206

Duane K. Miller, POC

The Lincoln Electric Company

22801 St. Clair Ave.

Cleveland, OH 44117-1194

Robert Pyle*

AISC Marketing

10101 South State Street

Sandy, Utah 84070

Douglas Rees-Evans*

Steel Dynamics, Inc.

Structural Mill Division

2601 County Road 700 East

Columbia City, IN 46725

Richard I. Seals

P.O. Box 11327

Berkeley, CA 94712-2327

Robert E. Shaw, Guideline Writer

Steel Structures Technology Center, Inc.

42400 W Nine Mile Road

Novi, MI 48375-4132

TAP for Connection Performance 

Charlie Carter*

American Institute of Steel Construction

One East Wacker Drive, Suite 3100

Chicago, IL 60601-2001

Robert H. Dodds

University of Illinois at Urbana-Champaign

205 N. Mathews Ave.

2129 Newmark Lab

Urbana, IL 61801

Roger Ferch, POC

Herrick Corporation

7021 Koll Center Parkway

P.O Box 9125

Pleasanton, CA 94566-9125

John D. Hooper, Guideline Writer

Skilling Ward Magnusson Barkshire, Inc.

1301 Fifth Avenue, Suite 3200

Seattle, WA 98101-2699

Egor Popov

University of California at Berkeley

Department of Civil and Environmental

Engineering, Davis Hall 

Berkeley, CA 94720 

Steve Powell*

SME Steel Contractors

5955 W. Wells Park Rd.

West Jordan, UT 84088

Charles W. Roeder, Team Leader

University of Washington

233-B More Hall FX-10

Dept. of Building and Safety

Seattle, WA 98195-2700

Stanley T. Rolfe

University of Kansas

Civil Engineering Department

2006 Learned Hall

Lawrence, KS 66045-2225

Rick Wilkinson*

Gayle Manufacturing Company

1455 East Kentucky

Woodland, CA 95695

S-5
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

SAC Project Participants Moment-Frame Buildings 

TAP for System Performance 

Jacques Cattan*

American Institute of Steel Construction

One East Wacker Drive, Suite 3100

Chicago, IL 60601-2001

Gary C. Hart

Hart Consultant Group

The Water Garden, Ste. 670E

2425 Olympic Blvd.

Santa Monica, CA 90404-4030

Y. Henry Huang*

Los Angeles County Dept. of Public Works

900 S. Fremont Avenue, 8th Floor

Alhambra, CA 91803

Helmut Krawinkler, Team Leader

Department of Civil Engineering

Stanford University

Stanford, CA 94305

Dennis Randall*

SME Steel Contractors

5955 West Wells Park Road

West Jordan, UT 84088

Andrei M. Reinhorn

State University of New York at Buffalo

Civil Engineering Department

231 Ketter Hall

Buffalo, NY 14260

Arthur E. Ross, JVMC

Cole/Yee/Shubert & Associates

2500 Venture Oaks Way, Suite 100

Sacramento, CA 95833

C. Mark Saunders, Guideline Writer

Rutherford & Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

W. Lee Shoemaker*

Metal Building Manufacturers Association

1300 Summer Avenue

Cleveland, OH 44115

John Theiss, POC

EQE/Theiss Engineers

1848 Lackland Hills Parkway

St. Louis, MO 63146-3572

TAP for Performance Prediction and Evaluation 

Vitelmo V. Bertero

University of California at Berkeley

Pacific Earthquake Engr. Research Center

1301 S. 46th St.

Richmond, CA 94804

Bruce R. Ellingwood

Johns Hopkins University

Department of Civil Engineering

3400 N. Charles St.

Baltimore, MD 21218

Douglas A. Foutch, Team Leader

University of Illinois

MC-250, 205 N. Mathews Ave.

3129 Newmark Civil Engineering Lab

Urbana, IL 61801

Theodore V. Galambos, POC

University of Minnesota

122 CE Building, 500 Pillsbury Dr. SE

Minnneapolis, MN 55455

Lawrence G. Griffis

Walter P. Moore & Associates

3131 Eastside, Second Floor

Houston, TX 77098

Edwin T. Huston, JVMC

Smith & Huston, Inc.

8618 Roosevelt Way NE

Seattle, WA 98115

S-6
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings SAC Project Participants 

Harry Martin*

American Iron and Steel Institute

11899 Edgewood Road, Suite G

Auburn, CA 95603

Thomas A. Sabol, Guideline Writer

Englekirk & Sabol Consulting Engineers

Tom Schlafly*

American Institute of Steel Construction

One East Wacker Drive, Suite 3100

Chicago, IL 60601-2001

P.O. Box 77-D

Los Angeles, CA 90007

NormAbrahamson

Pacific Gas & Electric

P.O. Box 770000, MC N4C

San Francisco, CA 94177

C.B. Crouse

URS – Dames and Moore

2025 First Avenue, Suite 500

Seattle, WA 98121

Technical Advisors 

Robert Kennedy

RPK Structural Mechanics Consultants

18971 Villa Terr

Yorba Linda, CA 92886

Social Economic and Policy Panel 

Martha Cox-Nitikman

Building and Owners and Managers

Association, Los Angeles 

700 South Flower, Suite 2325 

Los Angeles, CA 90017 

Karl Deppe

27502 Fawnskin Dr.

Rancho Palos Verdes, CA 90275

Eugene Lecomte

Institute for Business and Home Safety

6 Sheffield Drive

Billerica, MA 01821

James Madison

Attorney at Law, Mediator and Arbitrator

750 Menlo Avenue, Suite 250

Menlo Park, CA 94025

Alan Merson

Morley Builders

2901 28th Street, Suite 100

Santa Monica, CA 90405

Joanne Nigg

University of Delaware

Disaster Research Center

Newark, DE 19716

William Petak

University of Southern California

Lewis Hall, Room 201

650 Childs Way

Los Angeles, CA 90089

Francine Rabinovitz

Hamilton, Rabinovitz and Alschuler

1990 South Bundy Drive, Suite 777

Los Angeles, CA 90025

S-7
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

SAC Project Participants Moment-Frame Buildings 

Dennis Randall

SME Steel Contractors

5955 West Wells Park Road

West Jordan, UT 84088

David Ratterman

Stites and Harbison

400 West Market St., Suite 1800

Louisville, KY 40202-3352

L. Thomas Tobin, Panel Coordinator

134 California Ave.

Mill Valley, CA 94941

Stephen Toth

TIAA-CREF

730 Third Avenue

New York, NY 10017-3206

John H. Wiggins, POC

J.H. Wiggins Company

1650 South Pacific Coast Hwy, Suite 311

Redondo Beach, CA 90277

Performance of Steel Buildings in Past Earthquakes Subcontractors 

David Bonowitz

887 Bush, No. 610

San Francisco, CA 94108

Peter Clark

SAC Steel Project Technical Office

1301 South 46th St.

Richmond, CA 94804

Michael Durkin

Michael Durkin & Associates

22955 Leanora Dr.

Woodland Hills, CA 91367

James Goltz

California Institute of Technology

Office of Earthquake Programs

Mail Code 252-21

Pasadena, CA 91125

Bruce Maison

7309 Lynn Ave

Elcerrito, CA 94530

Peter Maranian

Brandow & Johnston Associates

1660 West Third Street

Los Angeles, CA 90017

Terrence Paret

Wiss Janney Elstner Associates, Inc.

2200 Powell St. Suite 925

Emeryville, CA 94602

Maryann Phipps

Degenkolb Engineers

225 Bush Street, Suite 1000

San Francisco, CA 94104

Allan Porush

Dames & Moore

911 Wilshire Blvd., Suite 700

Los Angeles, CA 90017

Access Current Knowledge Subcontractors 

David Bonowitz Stephen Liu

887 Bush , No. 610 Colorado School of Mines

San Francisco, CA 94108 Mathematics and Computer Science

Department 

Golden, CO 80401 

S-8
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings SAC Project Participants 

Materials and Fracture Subcontractors 

Robert Dexter Karl H. Frank

University of Minnesota University of Texas at Austin

122 Civil Engineering Building 10100 Bornet Rd.

500 Pillsbury Drive SE Ferguson Lab, P.R.C. #177

Minneapolis, MN 55455-0116 Austin, TX 78758

Welding and Inspection Subcontractors 

Pingsha Dong / Tom Kilinski Glenn M. Light / George Gruber 

Center for Welded Structures Research Southwest Research Institute 

Battelle Memorial Institute

501 King Avenue

Columbus, OH 43201-2693

Matthew Johnson

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

6220 Culebra Road, P. O. Drawer 28510

San Antonio, TX 78228-0510

William C. Mohr

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

Connection Performance Subcontractors 

Gregory Deierlein

Stanford University

Terman Engineering Center

Department of Civil and Enviromental Engr.

Stanford, CA 94305-4020

Charles W. Roeder

University of Washington

233-B More Hall FX-10

Seattle, WA 98195-2700

Sherif El-Tawil / Sashi Kunnath

University of Central Florida

Civil and Environmental Engr. Department

Orlando, FL. 32816-2450

Anthony Ingraffea

Cornell University

School of Civil Engineering

363 Hollister Hall

Ithaca, NY 14853

System Performance Subcontractors 

Paul Somerville

Woodward-Clyde Federal Services

566 El Dorado St., Suite 100

Pasadena, CA 91101-2560

Farzad Naeim

John A. Martin & Associates

1212 S. Flower Ave.

Los Angeles, CA 90015

Andrei M. Reinhorn

State University of New York at Buffalo

Civil Engineering Department

231 Ketter Hall

Buffalo, NY 14260

C. Allin Cornell

Stanford University

Terman Engineering Center

Stanford, CA 94305-4020

S-9
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

SAC Project Participants Moment-Frame Buildings 

Helmut Krawinkler

Dept. of Civil Engineering

Stanford University

Stanford, CA 94305

Gregory MacRae

University of Washington

Civil Engineering Department

Seattle, WA 98195-2700

Kazuhiko Kasai

Tokyo Institute of Technology

Structural Engineering Research Center

Nagatsuta, Midori-Ku

Yokohama 226-8503, JAPAN

Bruce F. Maison

7309 Lynn Avenue

El Cerrito, CA 94530

Performance Prediction and Evaluation Subcontractors 

James Anderson

University of Southern California

Civil Engineering Department

Los Angeles, CA 90089-2531

Douglas A. Foutch

University of Illinois

MC-250, 205 N. Mathews Ave.

3129 Newmark Civil Engineering Lab

Urbana, IL 61801

Gary C. Hart

Department of Civil and Environmental

Engineering 

University of California 

Los Angeles, CA 90095 

Y.K. Wen

University of Illinois

3129 Newmark Civil Engineering Lab

205 N. Mathews Ave.

Urbana, IL 61801

Testing Subcontractors 

Subhash Goel / Bozidar Stojadinovic

University of Michigan

Civil Engineering Department

Ann Arbor, MI 48109

Roberto Leon

Georgia Institute of Technology

School of Civil & Environmental Engr.

790 Atlantic Ave.

Atlanta, GA 30332-0355

Vitelmo V. Bertero / Andrew Whittaker

UC Berkeley

Pacific Earthquake Engr. Research Center

1301 S. 46th St.

Richmond, CA 94804

Thomas Murray

Virginia Tech, Dept. of Civil Engineering

200 Patton Hall

Blacksburg, VA 24061

James M. Ricles / Le-Wu Lu

Lehigh University

c/o ATLSS Center

117 ATLSS Drive, H Building

Bethlehem, PA 18015-4729

John M. Barsom

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

S-10
Recommended Seismic Design 

Criteria for New Steel FEMA-350 

Moment-Frame Buildings SAC Project Participants 

Hassan Astaneh

University of California at Berkeley

Dept. of Civil and Environmental Engr.

781 Davis Hall

Berkeley, CA 94720

Michael Engelhardt

University of Texas at Austin

Ferguson Laboratory

10100 Burnet Road, Building 177

Austin, TX 78712-1076

Gary T. Fry

Texas A&M University

Department of Civil Engineering

Constructed Facilities Division, CE/TTI

Building, Room 710D 

College Station, TX 77843-3136 

Chia-Ming Uang

University of California at San Diego

Dept. of AMES, Division of Structural Engr.

409 University Center

La Jolla, California 92093-0085

Stephen Schneider

University of Ilinois at Urbana-Champaign

3106 Newmark Civil Engr. Lab, MC-250

205 N. Mathews Avenue

Urbana, IL 61801

Matthew Johnson

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

James Anderson

University of Southern California

Civil Engineering Department

Los Angeles, CA 90089-2531

Bozidar Stojadinovic

Dept. of Civil & Environmental Engr.

University of California

Berkeley, CA 94720

Inspection Procedure Consultants 

Thomas Albert

Digiray Corporation

2235 Omega Road, No. 3

San Ramon, CA 94583

Randal Fong

Automated Inspection Systems, Inc.

4861 Sunrise Drive, Suite 101

Martinez, CA 94553

Andre Lamarre

R.D Tech, Inc.

1200 St. Jean Baptiste, Suite 120

Quebec City, Quebec, Canada G2ZE 5E8

Glenn Light

Southwest Research Institute

6220 Culebra Road

San Antonio, TX 78228

Andrey Mishin

AS & E High Energy Systems

330 Keller Street, Building 101

Santa Clara, CA 95054

Robert Shaw

Steel Structures Technology Center, Inc.

42400 W. Nine Mile Road

Novi, MI 48375-4132

Carlos Ventura

Dept of Civil Engineering

University of British Columbia

2324 Main Hall

Vancouver, BC, Canada V6T 1Z4

S-11
Recommended Seismic Design 

FEMA-350 Criteria for New Steel 

SAC Project Participants Moment-Frame Buildings 

Guideline Trial Applications Subcontractors 

John Hopper Lawrence Novak

Skilling Ward Magnusson Barkshire, Inc. Skidmore, Owings, and Merrill

1301 Fifth Avenue, Suite 320 

Seattle WA 98101-2699 

Leonard Joseph 

Thornton-Tomassetti Engineers 

641 6th Avenue, 7th Floor 

New York, NY 10011 

224 S. Michigan Ave, Suite 1000 

Chicago, IL 60604 

Maryann Phipps 

Degenkolb Engineers 

225 Bush Street, Suite 1000 

San Francisco, CA 94104 

Economic and Social Impact Study Subcontractors 

Ronald Eguchi

EQE Engineering and Design

300 Commerce Dr., Ste. 200

Irvine, CA 92602

Martin Gordon / Peter Morris

Adamson Associates

170 Columbus Avenue

San Francisco, CA 94133

Richard Henige

Lemessurier Consultants Inc.

675 Massachusetts Ave.

Cambridge, MA 02139-3309

Charles Kircher

Charles Kircher & Associates

1121 San Antonio Road, Suite D-202

Palo Alto, CA 94303

Lizandro Mercado

Brandow & Johnston Associates

1600 West 3rd St.

Los Angeles, CA 90017

Greg Schindler

KPFF Consulting Engineers

1201 3rd Ave.

Seattle, WA 98101-3000

Report Production and Administrative Services 

A. Gerald Brady, Technical Editor Carol Cameron, Publications Coordinator

Patricia A. Mork, Administrative Asst. Ericka Holmon, Admin. Assistant

Peter N. Mork, Computer Specialist California Universities for Research in

Bernadette A. Mosby, Operations Admin. Earthquake Engineering

Michelle S. Schwartzbach, Pub. Specialist 1301 S. 46th Street

Applied Technology Council Richmond, CA 94804

555 Twin Dolphin Drive, Suite 550

Redwood City, CA 94065

*indicates industrial or organizational contact representative 

S-12