DISCLAIMER 

This document provides recommended criteria for the seismic evaluation and upgrade of welded 

steel moment-frame buildings. The recommendations were developed by practicing engineers 

based on professional judgment and experience and supported by a large program of laboratory, 

field, and analytical research. While every effort has been made to solicit comments from a broad 

selection of the affected parties, this is not a consensus document. No warranty is offered, with 

regard to the recommendations contained herein, either by the Federal Emergency 

Management Agency, the SAC Joint Venture, the individual Joint Venture partners, or their 

directors, members or employees. These organizations and their employees do not assume 

any legal liability or responsibility for the accuracy, completeness, or usefulness of any of the 

information, products or processes included in this publication. The reader is cautioned to 

carefully review the material presented herein and exercise independent judgment as to its 

suitability for application to specific engineering projects. These recommended criteria have 

been prepared by the SAC Joint Venture with funding provided by the Federal Emergency 

Management Agency, under contract number EMW-95-C-4770. 

Cover Art. The beam-column connection assembly shown on the cover depicts the standard 

detailing used in welded steel moment-frame construction prior to the 1994 Northridge 

earthquake. This connection detail was routinely specified by designers in the period 1970-1994 

and was prescribed by the Uniform Building Code for seismic applications during the period 

1985-1994. It is no longer considered to be an acceptable design for seismic applications. 

Following the Northridge earthquake, it was discovered that many of these beam-column 

connections had experienced brittle fractures at the joints between the beam flanges and column 

flanges. 
Recommended Seismic Evaluation and Upgrade Criteria for 

Existing Welded Steel Moment-Frame Buildings 

SAC Joint Venture 

A partnership of

Structural Engineers Association of California (SEAOC)

Applied Technology Council (ATC)

California Universities for Research in Earthquake Engineering (CUREe)

Prepared for SAC Joint Venture Partnership by

Guidelines Development Committee

Ronald O. Hamburger, Chair 

John D. Hooper Thomas Sabol 

Robert Shaw C. Mark Saunders 

Lawrence D. Reaveley Raymond H. R. Tide 

Project Oversight Committee 

William J. Hall, Chair 

Shirin Ader

John M. Barsom

Roger Ferch

Theodore V. Galambos

John Gross

James R. Harris

Richard Holguin

Nestor Iwankiw 

Roy G. Johnston 

Len Joseph 

Duane K. Miller 

John Theiss 

John H. Wiggins 

SAC Project Management Committee 

SEAOC: William T. Holmes Program Manager: Stephen A. Mahin 

ATC: Christopher Rojahn 

CUREe: Robin Shepherd 

Project Director for Topical Investigations: 

James O. Malley 

Project Director for Product Development: 

Ronald O. Hamburger 

SAC Joint Venture 

SEAOC: www.seaoc.org 

ATC: www.atcouncil.org 

CUREe: www.curee.org 

June, 2000 
THE SAC JOINT VENTURE 

SAC is a joint venture of the Structural Engineers Association of California (SEAOC), the 

Applied Technology Council (ATC), and California Universities for Research in Earthquake Engineering 

(CUREe), formed specifically to address both immediate and long-term needs related to solving 

performance problems with welded steel moment-frame connections discovered following the 1994 

Northridge earthquake. SEAOC is a professional organization composed of more than 3,000 practicing 

structural engineers in California. The volunteer efforts of SEAOC’s members on various technical 

committees have been instrumental in the development of the earthquake design provisions contained in 

the Uniform Building Code as well as the National Earthquake Hazards Reduction Program (NEHRP) 

Recommended Provisions for Seismic Regulations for New Buildings and Other Structures. ATC is a 

nonprofit corporation founded to develop structural engineering resources and applications to mitigate 

the effects of natural and other hazards on the built environment. Since its inception in the early 1970s, 

ATC has developed the technical basis for the current model national seismic design codes for buildings; 

the de facto national standard for postearthquake safety evaluation of buildings; nationally applicable 

guidelines and procedures for the identification, evaluation, and rehabilitation of seismically hazardous 

buildings; and other widely used procedures and data to improve structural engineering practice. CUREe 

is a nonprofit organization formed to promote and conduct research and educational activities related to 

earthquake hazard mitigation. CUREe’s eight institutional members are: the California Institute of 

Technology, Stanford University, the University of California at Berkeley, the University of California at 

Davis, the University of California at Irvine, the University of California at Los Angeles, the University 

of California at San Diego, and the University of Southern California. These university earthquake 

research laboratory, library, computer and faculty resources are among the most extensive in the United 

States. The SAC Joint Venture allows these three organizations to combine their extensive and unique 

resources, augmented by consultants and subcontractor universities and organizations from around the 

nation, into an integrated team of practitioners and researchers, uniquely qualified to solve problems 

related to the seismic performance of steel moment-frame structures. 

ACKNOWLEDGEMENTS 

Funding for Phases I and II of the SAC Steel Program to Reduce the Earthquake Hazards of Steel 

Moment-Frame Structures was principally provided by the Federal Emergency Management Agency, 

with ten percent of the Phase I program funded by the State of California, Office of Emergency Services. 

Substantial additional support, in the form of donated materials, services, and data has been provided by 

a number of individual consulting engineers, inspectors, researchers, fabricators, materials suppliers and 

industry groups. Special efforts have been made to maintain a liaison with the engineering profession, 

researchers, the steel industry, fabricators, code-writing organizations and model code groups, building 

officials, insurance and risk-management groups, and federal and state agencies active in earthquake 

hazard mitigation efforts. SAC wishes to acknowledge the support and participation of each of the above 

groups, organizations and individuals. In particular, we wish to acknowledge the contributions provided 

by the American Institute of Steel Construction, the Lincoln Electric Company, the National Institute of 

Standards and Technology, the National Science Foundation, and the Structural Shape Producers 

Council. SAC also takes this opportunity to acknowledge the efforts of the project participants – the 

managers, investigators, writers, and editorial and production staff – whose work has contributed to the 

development of these documents. Finally, SAC extends special acknowledgement to Mr. Michael 

Mahoney, FEMA Project Officer, and Dr. Robert Hanson, FEMA Technical Advisor, for their continued 

support and contribution to the success of this effort. 
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings Table of Contents 

TABLE OF CONTENTS 

LIST OF FIGURES ..............................................................................................................................ix

LIST OF TABLES................................................................................................................................xi

1 INTRODUCTION............................................................................................................ 1-1

1.1 Purpose................................................................................................................. 1-1

1.2 Intent .................................................................................................................... 1-2

1.3 Background .......................................................................................................... 1-3

1.4 Application......................................................................................................... 1-10

1.5 Overview of these Recommended Criteria ........................................................ 1-10

2 EVALUATION OVERVIEW.......................................................................................... 2-1

2.1 Scope.................................................................................................................... 2-1

2.2 Steel Moment-Frame Building Construction....................................................... 2-1

2.2.1 Introduction.............................................................................................. 2-1

2.2.2 Welded Steel Moment-Frame (WSMF) Construction............................. 2-2

2.2.3 Damage to Welded Steel Moment-Frame (WSMF) Construction

in the 1994 Northridge, California Earthquake........................................ 2-4

2.2.4 Damage to Welded Steel Moment-Frame (WSMF) Construction

in Other Earthquakes................................................................................ 2-5

2.2.5 Post-Northridge Earthquake Construction Practice ................................. 2-6

2.3 Typical Pre-Northridge Connection Damage....................................................... 2-9

2.3.1 Girder Damage....................................................................................... 2-10

2.3.2 Column Flange Damage......................................................................... 2-12

2.3.3 Weld Damage, Defects, and Discontinuities ......................................... 2-14

2.3.4 Shear Tab Damage ................................................................................. 2-16

2.3.5 Panel Zone Damage ............................................................................... 2-17

2.3.6 Other Damage ........................................................................................ 2-19

2.4 Evaluation Procedures ....................................................................................... 2-19

2.5 Material Properties and Condition Assessments................................................ 2-21

2.5.1 Material Properties................................................................................. 2-21

2.5.2 Component Properties............................................................................ 2-26

2.5.3 Condition Assessment............................................................................ 2-27

2.5.3.1 Scope and Procedures ............................................................. 2-28

2.5.3.2 Quantifying Results................................................................. 2-29

3 PERFORMANCE EVALUATION ................................................................................. 3-1

3.1 Scope.................................................................................................................... 3-1

3.2 Performance Definitions ...................................................................................... 3-1

3.2.1 Hazards .................................................................................................... 3-3

3.2.1.1 General...................................................................................... 3-3

3.2.1.2 Ground Shaking ........................................................................ 3-3

3.2.1.3 Other Hazards ........................................................................... 3-5

3.2.2 Performance Levels.................................................................................. 3-5

3.2.2.1 Nonstructural Performance Levels............................................ 3-7

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Table of Contents Steel Moment-Frame Buildings 

3.2.2.2 Structural Performance Levels.................................................. 3-7

3.2.2.2.1 Collapse Prevention Performance Level ..................... 3-8

3.2.2.2.2 Immediate Occupancy Performance Level.................. 3-9

3.3 Evaluation Approach ......................................................................................... 3-10

3.4 Analysis.............................................................................................................. 3-11

3.4.1 Alternative Procedures........................................................................... 3-12

3.4.2 Procedure Selection ............................................................................... 3-13

3.4.3 Linear Static Procedure.......................................................................... 3-13

3.4.3.1 Basis of the Procedure............................................................. 3-13

3.4.3.2 Period Determination.............................................................. 3-16

3.4.3.3 Determination of Actions and Deformations.......................... 3-17

3.4.3.3.1 Pseudo Lateral Load................................................. 3-17

3.4.3.3.2 Vertical Distribution of Seismic Forces................... 3-19

3.4.3.3.3 Horizontal Distribution of Seismic Forces .............. 3-19

3.4.3.3.4 Diaphragms .............................................................. 3-19

3.4.3.3.5 Determination of Interstory Drift ............................. 3-19

3.4.3.3.6 Determination of Column Demands ........................ 3-20

3.4.4 Linear Dynamic Procedure..................................................................... 3-20

3.4.4.1 Basis of the Procedure............................................................. 3-20

3.4.4.2 Analysis................................................................................... 3-21

3.4.4.2.1 General ...................................................................... 3-21

3.4.4.2.2 Ground Motion Characterization............................... 3-22

3.4.4.3 Determination of Actions and Deformations.......................... 3-22

3.4.4.3.1 Factored Interstory Drift Demand ............................. 3-22

3.4.4.3.2 Determination of Column Demands.......................... 3-22

3.4.5 Nonlinear Static Procedure .................................................................... 3-22

3.4.5.1 Basis of the Procedure............................................................. 3-22

3.4.5.2 Analysis Considerations.......................................................... 3-23

3.4.5.2.1 General ...................................................................... 3-23

3.4.5.2.2 Control Node ............................................................. 3-24

3.4.5.2.3 Lateral Load Patterns................................................. 3-25

3.4.5.2.4 Period Determination ................................................ 3-25

3.4.5.2.5 Analysis of Three-Dimensional Models.................... 3-25

3.4.5.2.6 Analysis of Two-Dimensional Models...................... 3-25

3.4.5.2.7 Connection Modeling................................................ 3-25

3.4.5.3 Determination of Actions and Deformations.......................... 3-25

3.4.5.3.1 Target Displacement.................................................. 3-25

3.4.5.3.2 Diaphragms................................................................ 3-25

3.4.5.3.3 Factored Interstory Drift Demand ............................. 3-26

3.4.5.3.4 Multidirectional Effects............................................. 3-26

3.4.5.3.5 Factored Column and Column Splice Demands ....... 3-26

3.4.6 Nonlinear Dynamic Procedure............................................................... 3-26

3.4.6.1 Basis of the Procedure............................................................. 3-26

3.4.6.2 Analysis Assumptions............................................................. 3-27

3.4.6.2.1 General ...................................................................... 3-27

iv 
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Steel Moment-Frame Buildings Table of Contents 

3.4.6.2.2 Ground Motion Characterization............................... 3-27

3.4.6.3 Determination of Actions and Deformations.......................... 3-27

3.4.6.3.1 Response Quantities .................................................. 3-27

3.4.6.3.2 Factored Interstory Drift Demand ............................. 3-27

3.4.6.3.3 Factored Column and Column Splice Demands ....... 3-27

3.5 Mathematical Modeling..................................................................................... 3-28

3.5.1 Basic Assumptions................................................................................. 3-28

3.5.2 Frame Configuration.............................................................................. 3-28

3.5.2.1 Elements Modeled .................................................................. 3-28

3.5.2.2 Panel Zone Stiffness ............................................................... 3-29

3.5.3 Horizontal Torsion................................................................................. 3-29

3.5.4 Foundation Modeling............................................................................. 3-30

3.5.5 Diaphragms ............................................................................................ 3-30

3.5.6 P-D Effects............................................................................................. 3-31

3.5.7 Multidirectional Excitation Effects........................................................ 3-33

3.5.8 Vertical Ground Motion......................................................................... 3-33

3.6 Acceptance Criteria............................................................................................ 3-34

3.6.1 Factored-Demand-to-Capacity Ratio ..................................................... 3-34

3.6.2 Performance Limited by Interstory Drift Angle..................................... 3-37

3.6.2.1 Factored Interstory Drift Angle Demand ................................ 3-37

3.6.2.2 Factored Interstory Drift Angle Capacity................................ 3-38

3.6.2.2.1 Global Interstory Drift Angle .................................... 3-39

3.6.2.2.2 Local Interstory Drift Angle ...................................... 3-39

3.6.3 Performance Limited by Column Compressive Capacity...................... 3-40

3.6.3.1 Column Compressive Demand ............................................... 3-40

3.6.3.2 Column Compressive Capacity............................................... 3-42

3.6.4 Column Splice Capacity......................................................................... 3-42

3.6.4.1 Column Splice Tensile Demand ............................................. 3-42

3.6.4.2 Column Splice Tensile Capacity............................................. 3-42

4 LOSS ESTIMATION....................................................................................................... 4-1

4.1 Scope.................................................................................................................... 4-1

4.2 Loss Estimation Methods..................................................................................... 4-1

4.2.1 Use of Loss Estimation Methods ............................................................. 4-3

4.2.2 Scope of Loss Estimation Methods.......................................................... 4-4

4.3 Rapid Loss Estimation Method............................................................................ 4-4

4.3.1 Introduction.............................................................................................. 4-4

4.3.2 Seismic Demand Characterization........................................................... 4-5

4.3.3 Connection Damage Loss Functions........................................................ 4-7

4.3.4 Connection Restoration Cost Functions................................................. 4-10

4.3.5 Nonstructural Repair Cost Functions..................................................... 4-12

5 SEISMIC UPGRADE...................................................................................................... 5-1

5.1 Scope.................................................................................................................... 5-1

5.2 Codes and Standards............................................................................................ 5-1

5.3 Upgrade Objectives and Rehabilitation Criteria .................................................. 5-2

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Table of Contents Steel Moment-Frame Buildings 

5.3.1 Simplified Upgrade.................................................................................. 5-3

5.3.2 Systematic Upgrade ................................................................................. 5-4

5.4 Upgrade Strategies ............................................................................................... 5-6

5.4.1 Connection Modifications........................................................................ 5-7

5.4.2 Lessening or Removal of Irregularities.................................................... 5-8

5.4.3 Global Structural Stiffening..................................................................... 5-8

5.4.4 Global Structural Strengthening............................................................... 5-9

5.4.5 Mass Reduction...................................................................................... 5-10

5.4.6 Seismic Isolation.................................................................................... 5-11

5.4.7 Supplemental Energy Dissipation.......................................................... 5-12

5.5 As-Built Conditions ........................................................................................... 5-12

5.5.1 General................................................................................................... 5-12

5.5.2 Material and Section Properties ............................................................. 5-13

5.6 Upgrade Components......................................................................................... 5-13

5.6.1 Material Specifications .......................................................................... 5-13

5.6.2 Material Strength Properties .................................................................. 5-13

5.6.3 Mathematical Modeling......................................................................... 5-14

CONNECTION QUALIFICATION................................................................................ 6-1

6.1 Scope.................................................................................................................... 6-1

6.2 Performance Data for Existing Connections........................................................ 6-2

6.2.1 Welded Unreinforced Fully Restrained Connection................................ 6-2

6.2.1.1 Modeling Assumptions ............................................................. 6-4

6.2.1.1.1 Linear Analysis............................................................ 6-4

6.2.1.1.2 Nonlinear Analysis ...................................................... 6-4

6.2.1.2 Performance Qualification Data................................................ 6-4

6.2.2 Simple Shear Tab Connections – With Slabs .......................................... 6-4

6.2.2.1 Modeling Assumptions ............................................................. 6-6

6.2.2.1.1 Linear Analysis............................................................ 6-6

6.2.2.1.2 Nonlinear Analysis ...................................................... 6-7

6.2.2.2 Performance Qualification Data................................................ 6-7

6.2.3 Simple Shear Tab Connections –Without Slabs ...................................... 6-7

6.2.3.1 Modeling Assumptions ............................................................. 6-9

6.2.3.2 Performance Qualification Data................................................ 6-9

6.3 Basic Design Approach for Connection Upgrades .............................................. 6-9

6.3.1 Frame Configuration.............................................................................. 6-10

6.3.2 Required Drift Angle Capacity .............................................................. 6-12

6.3.3 Connection Configuration...................................................................... 6-14

6.3.4 Determine Plastic Hinge Locations........................................................ 6-14

6.3.5 Determine Probable Plastic Moment at Hinges ..................................... 6-15

6.3.6 Determine Shear at the Plastic Hinge .................................................... 6-16

6.3.7 Determine Strength Demands at Each Critical Section ......................... 6-16

6.3.8 Yield Moment ........................................................................................ 6-17

6.4 General Requirements........................................................................................ 6-18

6.4.1 Framing.................................................................................................. 6-18

6.4.1.1 Beam and Column Strength Ratio .......................................... 6-18

vi 

6 
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Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings Table of Contents 

6.4.1.2 Beam Flange Stability............................................................. 6-19

6.4.1.3 Beam Web Stability ................................................................ 6-20

6.4.1.4 Beam Span and Depth Effects ................................................ 6-20

6.4.1.5 Beam Flange Thickness Effects.............................................. 6-21

6.4.1.6 Lateral Bracing at Beam Flanges at Plastic Hinges ................ 6-22

6.4.1.7 Welded Shear Studs ................................................................ 6-22

6.4.2 Welded Joints......................................................................................... 6-23

6.4.2.1 Through Thickness Strength ................................................... 6-23

6.4.2.2 Base Material Toughness........................................................ 6-24

6.4.2.3 k-Area Properties .................................................................... 6-26

6.4.2.4 Weld Filler Metal Matching and Overmatching..................... 6-26

6.4.2.5 Weld Metal Toughness ........................................................... 6-27

6.4.2.6 Weld Backing, Weld Tabs and other Welding Details ........... 6-28

6.4.2.7 Reinforcing Fillet Welds and Weld Overlays ......................... 6-29

6.4.2.8 Weld Access Hole Size, Shape, Workmanship ...................... 6-29

6.4.2.9 Welding Quality Control and Quality Assurance ................... 6-30

6.4.3 Other Design Issues for Welded Connections ....................................... 6-31

6.4.3.1 Continuity Plates..................................................................... 6-31

6.4.3.2 Panel Zone Strength................................................................ 6-34

6.4.3.3 Connections to Column Minor Axis....................................... 6-35

6.4.3.4 Attachment of Other Construction.......................................... 6-35

6.4.4 Bolted Joint Requirements..................................................................... 6-36

6.4.4.1 Existing Conditions................................................................. 6-36

6.4.4.2 Connection Upgrades.............................................................. 6-36

6.5 Prequalified Connection Details – General........................................................ 6-36

6.5.1 Load Combinations and Resistance Factors .......................................... 6-37

6.6 Prequalified Connection Upgrades .................................................................... 6-37

6.6.1 Improved Welded Unreinforced Flange (IWURF) Connection............. 6-38

6.6.1.1 Design Procedure .................................................................... 6-41

6.6.2 Welded Bottom Haunch (WBH) Connection ........................................ 6-42

6.6.2.1 Design Procedure .................................................................... 6-42

6.6.3 Welded Top and Bottom Haunch (WTBH) Connection ....................... 6-45

6.6.3.1 Design Procedure .................................................................... 6-47

6.6.4 Welded Cover Plated Flange (WCPF) Connection................................ 6-47

6.6.4.1 Design Procedure .................................................................... 6-47

6.7 New Moment Frames and Moment-Resisting Connections .............................. 6-50

6.8 Proprietary Connections..................................................................................... 6-50

6.8.1 Side Plate (SP) Connection ................................................................... 6-52

6.8.2 Slotted Web (SW) Connection .............................................................. 6-53

6.8.3 Bolted Bracket (BB) Connection ........................................................... 6-55

6.9 Project-Specific Testing of Nonprequalified Connections ................................ 6-55

6.9.1 Testing Procedure .................................................................................. 6-56

6.9.2 Acceptance Criteria................................................................................ 6-59

6.9.3 Analytical Prediction of Behavior.......................................................... 6-60

6.10 Prequalification Testing Criteria........................................................................ 6-61

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6.10.1 Prequalification Testing......................................................................... 6-62

6.10.2 Extending the Limits on Prequalified Connections ............................... 6-62

APPENDIX A: DETAILED PROCEDURES FOR PERFORMANCE EVALUATION ......... A-1

A.1 Scope................................................................................................................... A-1

A.2 Performance Evaluation Approach ..................................................................... A-1

A.2.1 Performance Objectives and Confidence................................................ A-1

A.2.2 Basic Procedure....................................................................................... A-4

A.3 Determination of Hazard Parameters.................................................................. A-8

A.3.1 Spectral Response Acceleration.............................................................. A-8

A.3.2 Logarithmic Hazard Curve Slope ........................................................... A-8

A.4 Determination of Demand Factors.................................................................... A-10

A.5 Determination of Beam-Column Connection Assembly Capacities................. A-13

A.5.1 Connection Test Protocols .................................................................... A-13

A.5.2 Determination of Beam-Column Assembly Capacity and

Resistance Factors................................................................................. A-14

A.6 Global Stability Capacity.................................................................................. A-15

APPENDIX B: DETAILED PROCEDURES FOR LOSS ESTIMATION................................B-1

B.1 Introduction..........................................................................................................B-1

B.2 Scope....................................................................................................................B-2

B.2.1 General.....................................................................................................B-2

B.2.2 Typical Welded Steel Moment-Frame (WSMF) Buildings .....................B-3

B.3 Damage States......................................................................................................B-3

B.4 Basic Approach....................................................................................................B-5

B.5 Required Data – User Input .................................................................................B-7

B.5.1 Building Capacity Curve..........................................................................B-7

B.5.2 Structural Response ...............................................................................B-10

B.5.3 Structure Fragility ..................................................................................B-13

B.5.4 Loss Functions .......................................................................................B-16

B.6 Example Loss Estimates ....................................................................................B-18

REFERENCES, BIBLIOGRAPHY, AND ACRONYMS...........................................................R-1

SAC PROJECT PARTICIPANTS...............................................................................................S-1

viii 
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Steel Moment-Frame Buildings List of Figures 

LIST OF FIGURES

Figure 1-1 Typical Welded Moment-Resisting Connection Prior to 1994............................ 1-4

Figure 1-2 Common Zone of Fracture Initiation in Beam-Column Connection ................... 1-5

Figure 1-3 Fractures of Beam to Column Joints.................................................................... 1-5

Figure 1-4 Column Fractures................................................................................................. 1-6

Figure 1-5 Vertical Fracture through Beam and Shear Plate Connection ............................. 1-6

Figure 2-1 Typical Early Beam-Column Connections .......................................................... 2-1

Figure 2-2 Typical Bolted Web, Welded Flange Connection................................................ 2-3

Figure 2-3 Typical Cover Plated Connection ........................................................................ 2-9

Figure 2-4 Components of Moment Connection................................................................. 2-10

Figure 2-5 Types of Girder Damage.................................................................................... 2-11

Figure 2-6 Types of Column Damage ................................................................................. 2-13

Figure 2-7 Types of Weld Damage...................................................................................... 2-15

Figure 2-8 Types of Shear Tab Damage .............................................................................. 2-17

Figure 2-9 Types of Panel Zone Damage ............................................................................ 2-18

Figure 4-1 Connection Damage Ratio vs Modified Mercalli Intensity (MMI) ..................... 4-9

Figure 4-2 Connection Damage Ratio vs Peak Ground Acceleration (PGA)........................ 4-9

Figure 4-3 Connection Damage Ratio vs Building Pseudo Interstory Drift Ratio ............. 4-10

Figure 4-4 Connection Restoration Cost vs Modified Mercalli Intensity (MMI)................ 4-11

Figure 4-5 Connection Restoration Cost vs Peak Ground Acceleration (PGA).................. 4-11

Figure 4-6 Connection Restoration Cost vs Building Pseudo Interstory Drift Ratio ......... 4-12

Figure 4-7 Nonstructural Repair Cost vs Modified Mercalli Intensity (MMI).................... 4-13

Figure 4-8 Nonstructural Repair Cost vs Peak Ground Acceleration (PGA) ...................... 4-13

Figure 4-9 Nonstructural Repair Cost vs Building Pseudo Interstory Drift Ratio .............. 4-14

Figure 6-1 Welded Unreinforced Fully Restrained Connection (pre-1994).......................... 6-3

Figure 6-2 Typical Simple Shear Tab Connection with Slab ................................................ 6-6

Figure 6-3 Typical Simple Shear Tab Connection without Slab........................................... 6-8

Figure 6-4 Inelastic Behavior of Frames with Hinges in Beam Span.................................. 6-10

Figure 6-5 Location of Plastic Hinge Formation ................................................................. 6-15

Figure 6-6 Sample Calculation of Shear at Plastic Hinge ................................................... 6-17

Figure 6-7 Calculation of Demands at Critical Sections ..................................................... 6-17

Figure 6-8 Recommended Weld Access Hole Detail .......................................................... 6-31

Figure 6-9 Typical Continuity and Doubler Plates.............................................................. 6-33

Figure 6-10 Improved Welded Unreinforced Flange Connection ......................................... 6-39

Figure 6-11 Welding Requirements at Improved Welded Unreinforced

Flange Connection ............................................................................................. 6-40

Figure 6-12 Welded Bottom Haunch (WBH) Connection .................................................... 6-43

Figure 6-13 Welded Top and Bottom Haunch (WTBH) Connection.................................... 6-45

Figure 6-14 Welded Cover Plated Flange (WCPF) Connection............................................ 6-48

Figure 6-15 Proprietary Side Plate Connection – Application to Existing Construction ...... 6-52

Figure 6-16 Proprietary Slotted Web Connection ................................................................. 6-55

Figure 6-17 Bolted Bracket Connection ................................................................................ 6-56

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Recommended Seismic Evaluation and Upgrade 

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List of Figures Steel Moment-Frame Buildings 

Figure 6-18 Drift Angle ......................................................................................................... 6-57

Figure A-1 Representative Incremental Dynamic Analysis Plots ....................................... A-17

Figure B-1 Flowchart of Detailed Loss Estimation................................................................B-6

Figure B-2 Example Development of Standard (HAZUS-Compatible) Capacity

Curve from a Normalized Pushover Curve..........................................................B-9

Figure B-3 Example Demand Spectrum Construction and Calculation of Peak

Response Point (D, A)........................................................................................B-11

Figure B-4 Demand and Capacity of Typical 9-Story WSMF Buildings – Ground

Shaking of 1⁄2 the Design Earthquake.................................................................B-19

Figure B-5 Demand and Capacity of Typical 9-Story WSMF Buildings – Design

Earthquake Ground Shaking ..............................................................................B-20

Figure B-6 Demand and Capacity of Typical 9-Story WSMF Buildings – Maximum

Considered Earthquake Ground Shaking...........................................................B-20

Figure B-7 Structural Fragility Curves – Typical 9-Story Los Angeles Buildings

with Post-Northridge Connection Conditions....................................................B-22

Figure B-8 Discrete Damage-State Probability Curves – Typical 9-Story Los Angeles

Buildings with Post-Northridge Connection Conditions ...................................B-22

Figure B-9 Mean Structural Loss Ratio Curves – Typical 9-Story WSMF

Los Angeles Buildings .......................................................................................B-23

Figure B-10 Mean Structural Loss Rate Curves – Typical 9-Story WSMF

Los Angeles Buildings .......................................................................................B-24

Figure B-11 Mean Loss of Function Curves – Typical 9-Story WSMF

Los Angeles Buildings .......................................................................................B-25

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Steel Moment-Frame Buildings List of Tables 

LIST OF TABLES

Table 2-1 Types of Girder Damage.................................................................................... 2-11

Table 2-2 Types of Column Damage ................................................................................. 2-13

Table 2-3 Types of Weld Damage, Defects and Discontinuities ....................................... 2-15

Table 2-4 Types of Shear Tab Damage .............................................................................. 2-16

Table 2-5 Types of Panel Zone Damage ............................................................................ 2-18

Table 2-6 Default Material Specifications for WSMF Buildings ...................................... 2-25

Table 2-7 Lower Bound and Expected Material Properties for Structural Steel

Shapes of Various Grades.................................................................................. 2-26

Table 3-1 Building Performance Levels............................................................................... 3-6

Table 3-2 Structural Performance Levels............................................................................. 3-8

Table 3-3 Analysis Procedure Selection Criteria ............................................................... 3-14

Table 3-4 Modification Factors C3 for Linear Static Procedure......................................... 3-19

Table 3-5 Performance Parameters Requiring Evaluation of Confidence ......................... 3-34

Table 3-6 Factored-Demand-to-Capacity Ratios l for Specific Confidence Levels and

Uncertainty bUT factors ...................................................................................... 3-35

Table 3-7 Recommended Minimum Confidence Levels.................................................... 3-36

Table 3-8 Interstory Drift Angle Analysis Uncertainty Factors, ga .................................... 3-37

Table 3-9 Interstory Drift Angle Demand Variability Factors g......................................... 3-38

Table 3-10 Global Interstory Drift Angle Capacity C and Resistance Factors f

for Regular Buildings......................................................................................... 3-39

Table 3-11 Uncertainty Coefficient bUT for Global Interstory Drift Evaluation .................. 3-40

Table 3-12 Uncertainty Coefficient bUT for Local Interstory Drift Evaluation .................... 3-40

Table 3-13 Analysis Uncertainty Factor ga and Total Uncertainty Coefficient bUT for

Evaluation of Column Compressive Demands.................................................. 3-41

Table 5-1 Applicable Codes, Standards and Guideline Documents..................................... 5-1

Table 6-1 Performance Qualification Data – Welded Fully Restrained

Connection (pre-1994) ......................................................................................... 6-5

Table 6-2 Performance Qualification Data – Shear Tab Connections with Slabs ............... 6-8

Table 6-3 Performance Qualification Data – Shear Tab Connections (No Slab)................. 6-9

Table 6-4 Design Coefficients for SMF and OMF Systems .............................................. 6-13

Table 6-5 Column Flange Through-Thickness Strength .................................................... 6-23

Table 6-6 Prequalified Welded Fully Restrained Connection Upgrade Details................. 6-38

Table 6-7 Prequalification Data for Improved Welded Unreinforced

Flange Connections............................................................................................ 6-41

Table 6-8 Prequalification Data for Welded Bottom Haunch (WBH) Connection............ 6-44

Table 6-9 Prequalification Data for Welded Top and Bottom Haunch

(WTBH) Connections ........................................................................................ 6-46

Table 6-10 Prequalification Data for Welded Cover Plated Flange Connections................ 6-49

Table 6-11 Performance Data for Prequalified Moment-Resisting Connections

for New Framing................................................................................................ 6-51

Table 6-12 Interstory Drift Angle Limits for Various Performance Levels ......................... 6-56

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Table 6-13 Numerical Values of qj and nj ............................................................................ 6-58

Table 6-14 Minimum Qualifying Total Interstory Drift Angle Capacities, qSD

and qU, for OMF and SMF Systems .................................................................. 6-59

Table A-1 Confidence Parameter, l, as a Function of Confidence Level,

Hazard Parameter k, and Uncertainty bUT ........................................................... A-7

Table A-2 Default Values of the Logarithmic Hazard Curve Slope k for

Probabilistic Ground Shaking Hazards............................................................... A-9

Table A-3 Default Logarithmic Uncertainty bDU for Various Analysis Methods.............. A-12

Table A-4 Default Bias Factors CB .................................................................................... A-12

Table A-5 Behavior States for Performance Evaluation of Connection Assemblies......... A-14

Table B-1 Connections in Typical WSMF Buildings in Three Regions...............................B-3

Table B-2 Descriptions of Structural Damage States ...........................................................B-4

Table B-3 General Guidance for Expected Loss Ratio and Building Condition

in Each Damage State ..........................................................................................B-4

Table B-4 Specific Guidance for Selection of Damage State Based on Connection

Damage ................................................................................................................B-5

Table B-5 Capacity Curve Properties of Typical Welded Steel Moment-Frame

Buildings............................................................................................................B-10

Table B-6 Values of the Degradation Factor k for Typical WSMF Buildings ...................B-12

Table B-7 Maximum Interstory Drift Values Defining Damage-State Thresholds

of Typical WSMF Buildings..............................................................................B-14

Table B-8 Structural Damage-State Variability b Factors of Typical WSMF

Buildings............................................................................................................B-15

Table B-9 Mean Structural Loss Ratios and Rates of Typical WSMF Buildings...............B-17

Table B-10 Cleanup and Construction Time and Loss-of-Function Multipliers

for Typical WSMF Buildings ............................................................................B-18

Table B-11 Summary of Peak Response – Typical 9-Story WSMF Buildings ....................B-21

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Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings Chapter 1: Introduction 

1. INTRODUCTION 

1.1 Purpose 

This report, FEMA-351 – Recommended Seismic Evaluation and Upgrade Criteria for 

Existing Welded Steel Moment-Frame Buildings has been developed by the SAC Joint Venture 

under contract to the Federal Emergency Management Agency (FEMA) to provide structural 

engineers with recommended criteria for evaluation of the probable performance of existing steel 

moment-frame buildings in future earthquakes and to provide a basis for updating and revision of 

evaluation and rehabilitation guidelines and standards. It is one of a series of companion 

publications addressing the issue of the seismic performance of steel moment-frame buildings. 

The set of companion publications includes: 

• FEMA-350 – Recommended Seismic Design Criteria for New Steel Moment-Frame 

Buildings. This publication provides recommended criteria, supplemental to FEMA-302 – 

1997 NEHRP Recommended Provisions for Seismic Regulations for New Buildings and 

Other Structures, for the design and construction of steel moment-frame buildings and 

provides alternative performance-based design criteria. 

• FEMA-351 – Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded 

Steel Moment-Frame Buildings. This publication provides recommended methods to 

evaluate the probable performance of existing steel moment-frame buildings in future 

earthquakes and to retrofit these buildings for improved performance. 

• FEMA-352 – Recommended Postearthquake Evaluation and Repair Criteria for Welded Steel 

Moment-Frame Buildings. This publication provides recommendations for performing 

postearthquake inspections to detect damage in steel moment-frame buildings following an 

earthquake, evaluating the damaged buildings to determine their safety in the postearthquake 

environment, and repairing damaged buildings. 

• FEMA-353 – Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications. This publication provides 

recommended specifications for the fabrication and erection of steel moment frames for 

seismic applications. The recommended design criteria contained in the other companion 

documents are based on the material and workmanship standards contained in this document, 

which also includes discussion of the basis for the quality control and quality assurance 

criteria contained in the recommended specifications. 

The information contained in these recommended evaluation and upgrade criteria, hereinafter 

referred to as Recommended Criteria, is presented in the form of specific recommendations for 

design and performance evaluation procedures together with supporting commentary explaining 

part of the basis for these recommendations. Detailed derivations and explanations of the basis 

for these design and evaluation recommendations may be found in a series of State of the Art 

Reports prepared in parallel with these Recommended Criteria. These reports include: 

• FEMA-355A – State of the Art Report on Base Metals and Fracture. This report summarizes 

current knowledge of the properties of structural steels commonly employed in building 

construction, and the production and service factors that affect these properties. 

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Chapter 1: Introduction Steel Moment-Frame Buildings 

• FEMA-355B – State of the Art Report on Welding and Inspection. This report summarizes 

current knowledge of the properties of structural welding commonly employed in building 

construction, the effect of various welding parameters on these properties, and the 

effectiveness of various inspection methodologies in characterizing the quality of welded 

construction. 

• FEMA-355C – State of the Art Report on Systems Performance of Steel Moment Frames 

Subject to Earthquake Ground Shaking. This report summarizes an extensive series of 

analytical investigations into the demands induced in steel moment-frame buildings designed 

to various criteria, when subjected to a range of different ground motions. The behavior of 

frames constructed with fully restrained, partially restrained and fracture-vulnerable 

connections is explored for a series of ground motions, including motion anticipated at nearfault and soft-soil sites. 

• FEMA-355D – State of the Art Report on Connection Performance. This report summarizes 

the current state of knowledge of the performance of different types of moment-resisting 

connections under large inelastic deformation demands. It includes information on fully 

restrained, partially restrained, and partial strength connections, both welded and bolted, 

based on laboratory and analytical investigations. 

• FEMA-355E – State of the Art Report on Past Performance of Steel Moment-Frame 

Buildings in Earthquakes. This report summarizes investigations of the performance of steel 

moment-frame buildings in past earthquakes, including the 1995 Kobe, 1994 Northridge, 

1992 Landers, 1992 Big Bear, 1989 Loma Prieta and 1971 San Fernando events. 

• FEMA-355F – State of the Art Report on Performance Prediction and Evaluation of Steel 

Moment-Frame Buildings. This report describes the results of investigations into the ability 

of various analytical techniques, commonly used in design, to predict the performance of 

steel moment-frame buildings subjected to earthquake ground motion. Also presented is the 

basis for performance-based evaluation procedures contained in the design criteria and 

guideline documents, FEMA-350, FEMA-351, and FEMA-352. 

In addition to the recommended design criteria and the State of the Art Reports, a companion 

document has been prepared for building owners, local community officials and other nontechnical audiences who need to understand this issue. A Policy Guide to Steel Moment-Frame 

Construction (FEMA-354), addresses the social, economic, and political issues related to the 

earthquake performance of steel moment-frame buildings. FEMA-354 also includes discussion 

of the relative costs and benefits of implementing the recommended criteria. 

1.2 Intent 

These recommended seismic evaluation and upgrade criteria are intended as a resource 

document for organizations engaged in developing and updating guidelines and standards for 

seismic evaluation and upgrade of steel moment-frame buildings. These criteria have been 

developed by professional engineers and researchers, based on the findings of a large multi-year 

program of investigation and research into the performance of steel moment-frame structures. 

Development of these recommended criteria was not subjected to a formal consensus review and 

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Steel Moment-Frame Buildings Chapter 1: Introduction 

approval process, nor was formal review or approval obtained from SEAOC’s technical 

committees. However, it did include broad external review by practicing engineers, researchers, 

fabricators, erectors, inspectors, building officials, and the producers of steel and welding 

consumables. In addition, two workshops were convened to obtain direct comment from these 

stakeholders on the proposed recommendations. 

1.3 Background 

For many years, the basic intent of the building code seismic provisions has been to provide 

buildings with an ability to withstand intense ground shaking without collapse, but potentially 

with some significant structural damage. In order to accomplish this, one of the basic principles 

inherent in modern code provisions is to encourage the use of building configurations, structural 

systems, materials and details that are capable of ductile behavior. A structure behaves in a 

ductile manner if it is capable of withstanding large inelastic deformations without significant 

degradation in strength, and without the development of instability and collapse. The design 

forces specified by building codes for particular structural systems are related to the amount of 

ductility the system is deemed to possess. Generally, structural systems with more ductility are 

designed for lower forces than less ductile systems, as ductile systems are deemed capable of 

resisting demands that are significantly greater than their elastic strength limit. Starting in the 

1960s, engineers began to regard welded steel moment-frame buildings as being among the most 

ductile systems contained in the building code. Many engineers believed that welded steel 

moment-frame buildings were essentially invulnerable to earthquake-induced structural damage 

and thought that should such damage occur, it would be limited to ductile yielding of members 

and connections. Earthquake-induced collapse was not believed possible. Partly as a result of 

this belief, many industrial, commercial and institutional structures employing welded steel 

moment-frame systems were constructed, particularly in the western United States. 

The Northridge earthquake of January 17, 1994 challenged this paradigm. Following that 

earthquake, a number of welded steel moment-frame buildings were found to have experienced 

brittle fractures of beam-to-column connections. The damaged buildings had heights ranging 

from one story to 26 stories, and a range of ages spanning from buildings as old as 30 years to 

structures being erected at the time of the earthquake. The damaged buildings were spread over a 

large geographical area, including sites that experienced only moderate levels of ground shaking. 

Although relatively few buildings were located on sites that experienced the strongest ground 

shaking, damage to buildings on these sites was extensive. Discovery of these unanticipated 

brittle fractures of framing connections, often with little associated architectural damage to the 

buildings, was alarming to engineers and the building industry. The discovery also caused some 

concern that similar, but undiscovered, damage may have occurred in other buildings affected by 

past earthquakes. Later investigations confirmed such damage in a limited number of buildings 

affected by the 1992 Landers, 1992 Big Bear and 1989 Loma Prieta earthquakes. 

In general, welded steel moment-frame buildings damaged by the 1994 Northridge 

earthquake met the basic intent of the building codes. That is, they experienced limited structural 

damage, but did not collapse. However, the structures did not behave as anticipated and 

significant economic losses occurred as a result of the connection damage, in some cases, in 

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Chapter 1: Introduction Steel Moment-Frame Buildings 

buildings that had experienced ground shaking less severe than the design level. These losses 

included direct costs associated with the investigation and repair of this damage as well as 

indirect losses relating to the temporary and, in a few cases, long-term loss of use of space within 

damaged buildings. 

Welded steel moment-frame buildings are anticipated to develop their ductility through the 

development of yielding in beam-column assemblies at the beam-column connections. This 

yielding may take the form of plastic hinging in the beams (or, less desirably, in the columns), 

plastic shear deformation in the column panel zones, or through a combination of these 

mechanisms. It was believed that the typical connection employed in welded steel momentframe construction, shown in Figure 1-1, was capable of developing large plastic rotations, on the 

order of 0.02 radians or larger, without significant strength degradation. 

Figure 1-1 Typical Welded Moment-Resisting Connection Prior to 1994 

Observation of damage sustained by buildings in the 1994 Northridge earthquake indicated 

that, contrary to the intended behavior, in many cases, brittle fractures initiated within the 

connections at very low levels of plastic demand, and in some cases, while the structures 

remained essentially elastic. Typically, but not always, fractures initiated at the complete joint 

penetration (CJP) weld between the beam bottom flange and column flange (Figure 1-2). Once 

initiated, these fractures progressed along a number of different paths, depending on the 

individual joint conditions. 

In some cases, the fractures progressed completely through the thickness of the weld, and 

when fire protective finishes were removed, the fractures were evident as a crack through 

exposed faces of the weld, or the metal just behind the weld (Figure 1-3a). Other fracture 

patterns also developed. In some cases, the fracture developed into a crack of the column flange 

material behind the CJP weld (Figure 1-3b). In these cases, a portion of the column flange 

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Recommended Seismic Evaluation and Upgrade 

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Steel Moment-Frame Buildings Chapter 1: Introduction 

remained bonded to the beam flange, but pulled free from the remainder of the column. This 

fracture pattern has sometimes been termed a “divot” or “nugget” failure. 

Backing bar Backing bar

Column flange Column flange

Beam flange Beam flange

Fused zone Fused zone

Fracture Fracture

Backing bar Backing bar

Column flange Column flange

Beam flange Beam flange

Fused zone Fused zone

Fracture Fracture

Figure 1-2 Common Zone of Fracture Initiation in Beam-Column Connection 

A number of fractures progressed completely through the column flange, along a nearhorizontal plane that aligns approximately with the beam lower flange (Figure 1-4a). In some 

cases, these fractures extended into the column web and progressed across the panel zone (Figure 

1-4b). Investigators have reported some instances where columns fractured entirely across the 

section. 

a. Fracture at Fused Zone b. Column Flange "Divot" Fracture 

Figure 1-3 Fractures of Beam to Column Joints 

Once such fractures occur, the beam-column connection loses a significant portion of the 

flexural rigidity and strength needed to resist loads that tend to open the crack. Residual flexural 

strength and rigidity must be developed through a couple consisting of forces transmitted through 

the remaining flange connection and the web bolts. However, in providing this residual strength 

and stiffness, the bolted web connections can themselves be subject to failures. These include 

fracturing of the welds of the shear plate to the column, fracturing of supplemental welds to the 

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Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

Chapter 1: Introduction Steel Moment-Frame Buildings 

beam web or fracturing through the weak section of shear plate aligning with the bolt holes 

(Figure 1-5). 

a. Fractures through Column Flange b. Fracture Progresses into Column Web 

Figure 1-4 Column Fractures 

Despite the obvious local strength impairment resulting from these fractures, many damaged 

buildings did not display overt signs of structural damage, such as permanent drifts or damage to 

architectural elements, making reliable postearthquake damage evaluations difficult. In order to 

determine reliably if a building has sustained connection damage it is necessary to remove 

architectural finishes and fireproofing, and perform detailed inspections of the connections. 

Even if no damage is found, this is a costly process. Repair of damaged connections is even 

more costly. At least one welded steel moment-frame building sustained so much damage that it 

was deemed more practical to demolish the building than to repair it. 

Figure 1-5 Vertical Fracture through Beam Shear Plate Connection 

Initially, the steel construction industry took the lead in investigating the causes of this 

unanticipated damage and in developing design recommendations. The American Institute of 

Steel Construction (AISC) convened a special task committee in March, 1994 to collect and 

disseminate available information on the extent of the problem (AISC, 1994a). In addition, 

together with a private party engaged in the construction of a major steel building at the time of 

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Steel Moment-Frame Buildings Chapter 1: Introduction 

the earthquake, AISC participated in sponsoring a limited series of tests of alternative connection 

details at the University of Texas at Austin (AISC, 1994b). The American Welding Society 

(AWS) also convened a special task group to investigate the extent that the damage related to 

welding practice and to determine if changes to the welding code were appropriate (AWS, 1995). 

In September, 1994, the SAC Joint Venture, AISC, the American Iron and Steel Institute and 

National Institute of Standards and Technology jointly convened an international workshop 

(SAC, 1994) in Los Angeles to coordinate the efforts of the various participants and to lay the 

foundation for systematic investigation and resolution of the problem. Following this workshop, 

FEMA entered into a cooperative agreement with the SAC Joint Venture to perform problemfocused studies of the seismic performance of steel moment-frame buildings and to develop 

recommendations for professional practice (Phase I of SAC Steel Project). Specifically, these 

recommendations were intended to address the following: the inspection of earthquake-affected 

buildings to determine if they had sustained significant damage; the repair of damaged buildings; 

the upgrade of existing buildings to improve their probable future performance; and the design of 

new structures to provide reliable seismic performance. 

During the first half of 1995, an intensive program of research was conducted to explore 

more definitively the pertinent issues. This research included literature surveys, data collection 

on affected structures, statistical evaluation of the collected data, analytical studies of damaged 

and undamaged buildings, and laboratory testing of a series of full-scale beam-column 

assemblies representing typical pre-Northridge design and construction practice as well as 

various repair, upgrade and alternative design details. The findings of these tasks formed the 

basis for the development of FEMA-267 – Interim Guidelines: Evaluation, Repair, Modification, 

and Design of Welded Steel Moment Frame Structures, which was published in August, 1995. 

FEMA-267 provided the first definitive, albeit interim, recommendations for practice, following 

the discovery of connection damage in the 1994 Northridge earthquake. 

In September 1995 the SAC Joint Venture entered into a contractual agreement with FEMA 

to conduct Phase II of the SAC Steel Project. Under Phase II, SAC continued its extensive 

problem-focused study of the performance of moment resisting steel frames and connections of 

various configurations, with the ultimate goal of develop seismic design criteria for steel 

construction. This work has included: extensive analyses of buildings; detailed finite element 

and fracture mechanics investigations of various connections to identify the effects of connection 

configuration, material strength, and toughness and weld joint quality on connection behavior; as 

well as more than 120 full-scale tests of connection assemblies. As a result of these studies, and 

independent research conducted by others, it is now known that the typical moment-resisting 

connection detail employed in steel moment-frame construction prior to the 1994 Northridge 

earthquake, and depicted in Figure 1-1, had a number of features that rendered it inherently 

susceptible to brittle fracture. These included the following: 

• The most severe stresses in the connection assembly occur where the beam joins to the 

column. Unfortunately, this is also the weakest location in the assembly. At this location, 

bending moments and shear forces in the beam must be transferred to the column through the 

combined action of the welded joints between the beam flanges and column flanges and the 

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FEMA-351 Criteria for Existing Welded 

Chapter 1: Introduction Steel Moment-Frame Buildings 

shear tab. The combined section properties of these elements, for example the cross sectional 

area and section modulus, are typically less than those of the connected beam. As a result, 

stresses are locally intensified at this location. 

• The joint between the bottom beam flange and the column flange is typically made as a 

downhand field weld, often by a welder sitting on top of the beam top flange, in a so-called 

“wildcat” position. To make the weld from this position each pass must be interrupted at the 

beam web, with either a start or stop of the weld at this location. This welding technique 

often results in poor quality welding at this critical location, with slag inclusions, lack of 

fusion and other defects. These defects can serve as crack initiators, when the connection is 

subjected to severe stress and strain demands. 

• The basic configuration of the connection makes it difficult to detect hidden defects at the 

root of the welded beam-flange-to-column-flange joints. The backing bar, which was 

typically left in place following weld completion, restricts visual observation of the weld 

root. Therefore, the primary method of detecting defects in these joints is through the use of 

ultrasonic testing (UT). However, the geometry of the connection also makes it very difficult 

for UT to detect flaws reliably at the bottom beam flange weld root, particularly at the center 

of the joint, at the beam web. As a result, many of these welded joints have undetected 

significant defects that can serve as crack initiators. 

• Although typical design models for this connection assume that nearly all beam flexural 

stresses are transmitted by the flanges and all beam shear forces by the web, in reality, due to 

boundary conditions imposed by column deformations, the beam flanges at the connection 

carry a significant amount of the beam shear. This results in significant flexural stresses on 

the beam flange at the face of the column, and also induces large secondary stresses in the 

welded joint. Some of the earliest investigations of these stress concentration effects in the 

welded joint were conducted by Richard, et al. (1995). The stress concentrations resulting 

from this effect resulted in severe strength demands at the root of the complete joint 

penetration welds between the beam flanges and column flanges, a region that often includes 

significant discontinuities and slag inclusions, which are ready crack initiators. 

• In order that the welding of the beam flanges to the column flanges be continuous across the 

thickness of the beam web, this detail incorporates weld access holes in the beam web, at the 

beam flanges. Depending on their geometry, severe strain concentrations can occur in the 

beam flange at the toe of these weld access holes. These strain concentrations can result in 

low-cycle fatigue and the initiation of ductile tearing of the beam flanges after only a few 

cycles of moderate plastic deformation. Under large plastic flexural demands, these ductile 

tears can quickly become unstable and propagate across the beam flange. 

• Steel material at the center of the beam-flange-to-column-flange joint is restrained from 

movement, particularly in connections of heavy sections with thick column flanges. This 

condition of restraint inhibits the development of yielding at this location, resulting in locally 

high stresses on the welded joint, which exacerbates the tendency to initiate fractures at 

defects in the welded joints. 

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Steel Moment-Frame Buildings Chapter 1: Introduction 

• Design practice in the period 1985-1994 encouraged design of these connections with 

relatively weak panel zones. In connections with excessively weak panel zones, inelastic 

behavior of the assembly is dominated by shear deformation of the panel zone. This panel 

zone shear deformation results in a local kinking of the column flanges adjacent to the beamflange-to-column-flange joint, and further increases the stress and strain demands in this 

sensitive region. 

In addition to the above, additional conditions contributed significantly to the vulnerability of 

connections constructed prior to 1994. 

• In the mid-1960s, the construction industry moved to the use of the semi-automatic, selfshielded, flux-cored arc welding process (FCAW-S) for making the joints of these 

connections. The welding consumables that building erectors most commonly used 

inherently produced welds with very low toughness. The toughness of this material could be 

further compromised by excessive deposition rates, which unfortunately were commonly 

employed by welders. As a result, brittle fractures could initiate in welds with large defects, 

at stresses approximating the yield strength of the beam steel, precluding the development of 

ductile behavior. 

• Early steel moment frames tended to be highly redundant and nearly every beam-column joint 

was constructed to behave as part of the lateral-force-resisting system. As a result, member 

sizes in these early frames were small and much of the early acceptance testing of this typical 

detail was conducted with specimens constructed of small framing members. As the cost of 

construction labor increased, the industry found that it was more economical to construct 

steel moment-frame buildings by moment-connecting a relatively small percentage of the 

beams and columns and by using larger members for these few moment-connected elements. 

The amount of strain demand placed on the connection elements of a steel moment frame is 

related to the span-to-depth ratio of the member. Therefore, as member sizes increased, 

strain demands on the welded connections also increased, making the connections more 

susceptible to brittle behavior. 

• In the 1960s and 1970s, when much of the initial research on steel moment-frame 

construction was performed, beams were commonly fabricated using A36 material. In the 

1980s, many steel mills adopted more modern production processes, including the use of 

scrap-based production. Steels produced by these more modern processes tended to include 

micro-alloying elements that increased the strength of the materials so that despite the 

common specification of A36 material for beams, many beams actually had yield strengths 

that approximated or exceeded that required for grade 50 material. As a result of this 

increase in base metal yield strength, the weld metal in the beam-flange-to-column-flange 

joints became under-matched, potentially contributing to its vulnerability. 

At this time, it is clear that in order to obtain reliable ductile behavior of steel moment-frame 

construction a number of changes to past practices in design, materials, fabrication, erection and 

quality assurance are necessary. The recommended criteria contained in this document, and the 

companion publications, are based on an extensive program of research into materials, welding 

technology, inspection methods, frame system behavior, and laboratory and analytical 

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Chapter 1: Introduction Steel Moment-Frame Buildings 

investigations of different connection details. The guidelines presented herein are believed to be 

capable of addressing the vulnerabilities identified above and providing for frames capable of 

more reliable performance in response to earthquake ground shaking. 

1.4 Application 

These Recommended Criteria supersede the evaluation and upgrade recommendations for 

existing WSMF buildings contained in FEMA-267, Interim Guidelines: Evaluation, Repair, 

Modification and Design of Welded Steel Moment Frame Structures, and the Interim Guidelines 

Advisories, Nos. 1 and 2 (FEMA-267A and FEMA-267B). It is intended to be used as a basis for 

updating and revision of evaluation and rehabilitation guidelines and standards currently 

employed in welded steel moment-frame construction in order to permit more reliable seismic 

performance. Some users may wish to apply these Recommended Criteria to specific 

engineering projects, prior to their adoption by future codes and standards. Such users are 

cautioned to consider carefully the codes and standards actually enforced by the building 

department having jurisdiction for a specific project, and to adjust the Recommended Criteria 

accordingly. These users are also cautioned that these recommendations have not undergone a 

consensus adoption process. Users should thoroughly acquaint themselves with the technical 

data upon which these recommendations are based and exercise their own independent 

engineering judgment prior to implementing them in practice. 

1.5 Overview of These Recommended Criteria 

The following is an overview of the general contents of the chapters contained in these 

Recommended Criteria, and their intended use: 

• Chapter 2: Evaluation Overview. This chapter provides an historic perspective of the 

development of steel moment-frame design and construction practice in the United States. It 

also includes discussion of the performance of welded steel moment-frame construction in 

recent earthquakes and the causes for much of the damage observed in this construction. 

Guidelines for collection of basic data on the configuration, and the details and materials of 

construction of a building, needed to conduct an evaluation, are presented, as is a brief 

introduction into the types of evaluation that may be conducted. 

• Chapter 3: Performance Evaluation. This chapter presents simplified analytical 

procedures for determining the probable structural performance of regular, welded, steel 

moment-frame buildings, given the site seismicity. These procedures allow the calculation of 

a level of confidence (say, 95%) that an existing structure will achieve a stipulated 

performance level (e.g., a Collapse Prevention level) for a specified earthquake hazard (e.g., a 

2% probability of exceedence in 50 years). If the calculated level of confidence is 

unacceptably low, then the structure can be upgraded and re-evaluated for more acceptable 

performance, using these same procedures. 

• Chapter 4: Loss Estimation. This chapter presents a simplified procedure for estimating 

the probable postearthquake repair costs for existing, welded, steel moment-frame buildings 

using basic information on the building’s configuration and age, and the intensity of ground 

shaking at the site. 

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Steel Moment-Frame Buildings Chapter 1: Introduction 

• Chapter 5: Seismic Upgrade. This chapter presents recommendations for two approaches 

to seismic upgrade of existing, welded, steel moment-frame buildings. The first approach, 

termed simplified upgrade, consists of modification of individual moment-resisting 

connections to reduce their susceptibility to ground-shaking-induced brittle fracture. The 

second method is a detailed procedure in which the performance of the structure is first 

evaluated, using the procedures of Chapter 3, an upgrade approach is conceived and designed 

in a preliminary manner, and the performance of the upgraded structure is evaluated for 

acceptability. This process is repeated until a suitable level of confidence of acceptable 

performance is obtained. Upgrades in this second method may consist of connection 

upgrades, as in the simplified upgrade approach, but may also include modification of the 

structural system, such as introduction of braces, or energy dissipation devices. 

• Chapter 6: Connection Qualification. This chapter presents modeling recommendations 

and performance data for different types of beam-column connections. 

• Appendix A: Detailed Procedures for Performance Evaluation. This appendix provides 

recommendations for the implementation of the detailed analytical performance evaluation 

procedures upon which the simplified procedures of Chapter 3 are based. Implementation of 

these procedures can permit more certain evaluation of the performance of a building to be 

determined than is possible using the simplified methods of Chapter 3. Engineers may find 

the application of these more detailed procedures beneficial in demonstrating that building 

performance is better than indicated by Chapter 3. Use of these more detailed procedures is 

required for the performance evaluation of structures with certain irregularities, as indicated 

in Chapter 3. 

• Appendix B: Detailed Procedures for Loss Estimation. This appendix provides 

procedures for developing building-specific, vulnerability (and loss) functions for steel 

moment-frame buildings. These vulnerability and loss functions are compatible with 

HAZUS, a nationally applicable computer program developed by FEMA that permits 

estimation of earthquake losses on a building-specific basis, or community or regional basis. 

These vulnerability and loss functions may also be used with other loss-modeling software 

and methodologies. 

• References, Bibliography, and Acronyms. 

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2. Evaluation Overview 

2.1 Scope 

This section provides a discussion of the history of the development of steel moment-frame 

buildings and the general earthquake damage and vulnerabilities associated with such buildings. 

An overview of the evaluation procedures contained in these recommended criteria is presented 

along with corresponding sections regarding material property and condition assessment 

approaches. 

2.2 Steel Moment-Frame Building Construction 

2.2.1 Introduction 

Steel frames have been used in building construction for more than one hundred years. In the 

early 20th century, typical steel frames were of riveted construction. Beam-column connections 

were of two common types illustrated in Figure 2-1, in which beams were connected to columns 

using either stiffened or unstiffened angles at the top and bottom beam flanges. Designers often 

assumed that these assemblies acted as “pinned” connections for gravity loads and that the 

stiffened connections would act as “fixed” connections for lateral loads. Although some hotrolled shapes were available, these were typically limited to beam applications. Columns and 

girders were often fabricated out of plate and angle sections. Frames were typically designed for 

lateral wind loading, employing approximate methods of frame analysis, such as the portal 

method or cantilever method. 

Figure 2-1 Typical Early Beam-Column Connections 

Most early steel frame buildings had exterior walls of unreinforced masonry. The exterior 

building frame was typically embedded in these walls providing for significant interaction 

between the steel and masonry elements. Although these buildings were usually designed 

neglecting the effects of the masonry in lateral load resistance, in actuality there is significant 

interaction between the masonry walls and steel frames and the masonry provides much of the 

lateral resistance of such buildings. 

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Infilled masonry construction remained common until the early 1940s. At about that time, 

reinforced, cast-in-place concrete walls began to replace the masonry used in earlier buildings. 

These reinforced concrete walls were typically designed to provide the lateral resistance for the 

structure, and the steel frame was often designed only to carry gravity loading, though some 

buildings with a “dual” system of concrete walls and steel moment frames also were constructed 

during this period. Steel moment frames without infill walls came into wider use when curtain 

wall systems became popular, in the late 1940s and early 1950s. This was the time when 

moment resistance and stiffness of the connections became a critical issue. The earliest steel 

moment frames employed riveted or bolted connections similar to those used in the earliest infill 

masonry buildings. However, as design procedures became more sophisticated and the building 

codes began to require design for larger seismic forces, designers started to design fully 

restrained connections intended to develop the full flexural strength of the beams. Connections 

were usually complex and expensive, consisting, for example, of plates, stiffened angles, and Tsections that were riveted or bolted. 

During the Second World War, structural welding was introduced in the ship-building 

industry as a means of speeding ship construction. It is interesting to note that these early 

attempts at welded construction were not entirely successful and were plagued by unanticipated 

fracture problems. Several Liberty Ships, a class of cargo vessel, some of which were among the 

first to employ welded hull construction, experienced massive fracture damage and a few actually 

fractured in two and sank. These problems were eventually traced to sharp corners at openings in 

the hull and superstructure as well as to inadequate notch toughness in the materials of 

construction. By the 1950s, however, these problems were largely mitigated by improved design 

and construction practice and welded construction had completely replaced the earlier bolted and 

riveted construction techniques formerly prevalent in this industry. 

In the late 1950s, structural welding began to spread to the building industry. This trend, 

together with the need to design strong and stiff, but economical, connections, accelerated a 

design shift from riveted or bolted, partially restrained connections to designs employing welded, 

fully restrained connections. Many different types of welded connections were used, the earlier 

ones consisting mostly of shop-welded, field-bolted cover plates connecting the beam flanges to 

the columns. In the late 1950s the field-welded direct connection between beam flanges and 

column flanges started to see some use. Experimental research performed in the mid to late 

1950s, primarily at Lehigh University, provided criteria for welding and for continuity plate 

requirements to minimize web crippling and column flange distortions. Additional experimental 

research performed in the mid 1960s to early 1970s at the University of California at Berkeley 

provided evidence that certain types of butt-welded beam-flange-to-column-flange connections 

could behave satisfactorily under cyclic loading. These data lead to widespread adoption of the 

bolted-web, welded-flange, beam-column connection shown in Figure 2-2, by engineers 

designing for earthquake resistance. 

2.2.2 Welded Steel Moment-Frame (WSMF) Construction 

Today, WSMF construction is commonly used throughout the United States and the world, 

particularly for mid-rise and high-rise construction. Prior to the 1994 Northridge earthquake, this 

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type of construction was considered one of the most seismic-resistant structural systems, due to 

the fact that severe damage to such structures had rarely been reported in past earthquakes and 

there was no record of earthquake-induced collapse of such buildings, constructed in accordance 

with contemporary US practice. However, the widespread reports of structural damage to such 

structures following the Northridge earthquake called for re-examination of this premise. 

Figure 2-2 Typical Bolted Web, Welded Flange Connection 

Steel moment-frame buildings are designed to resist earthquake ground shaking, based on the 

assumption that they are capable of extensive yielding and plastic deformation, without loss of 

strength. The intended plastic deformation consists of plastic rotations developing within the 

beams, at their connections to the columns, and is theoretically capable of resulting in benign 

dissipation of the earthquake energy delivered to the building. Damage is expected to consist of 

moderate yielding and localized buckling of the steel elements, but not brittle fractures. Based 

on this presumed behavior, building codes permit design of steel moment-frame structures for 

lateral forces that are approximately 1/8 those which would be required for the structure to 

remain fully elastic. Supplemental provisions within the building code, intended to control the 

amount of interstory drift sustained by these flexible frame buildings, typically result in 

structures which are substantially stronger than this minimum requirement and in zones of 

moderate seismicity, substantial overstrength may be present to accommodate wind and gravity 

load design conditions. In zones of high seismicity, most such structures designed to minimum 

code criteria will not start to exhibit plastic behavior until ground motions are experienced that 

are 1/3 to 1/2 the severity anticipated as a design basis. This design approach has been developed 

based on historical precedent, the observation of steel building performance in past earthquakes, 

limited research that has included laboratory testing of beam-column models (albeit with mixed 

results), and nonlinear analytical studies. 

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2.2.3 Damage to Welded Steel Moment-Frame (WSMF) Construction in the 1994 

Northridge, California, Earthquake 

Following the apparent widespread discovery of steel frame damage in the 1994 Northridge 

earthquake, the City of Los Angeles enacted an ordinance requiring mandatory inspections of 

approximately 240 buildings located in the zones of heaviest ground shaking within the City. 

This ordinance required that a report be filed for each building, indicating that inspections had 

been performed in accordance with FEMA 267, Interim Guidelines: Evaluation, Repair, 

Modification and Design of Welded Steel Moment Frame Structures, or other suitable approach, 

and that repairs be made. The resulting database of reported information provides a good 

overview of the types of damage sustained by buildings in the Northridge earthquake, though 

some damaged buildings, located in the zones of the most severe ground shaking, were outside 

the City of Los Angeles and were not included under the ordinance. 

Review of statistics obtained from a data base of the damage reported under this ordinance 

program indicates that the damage was less severe than had originally been perceived. Reports 

for approximately one third of the buildings affected by the ordinance indicated that no damage 

was found in the structures. Reports for another one third of the buildings indicated only that 

there were rejectable defects at the roots of some beam-flange-to-column-flange welds. At the 

time these inspections were made, there was some uncertainty as to whether such conditions 

were actually damage or poor quality construction, which had not been detected during the 

original performance of construction quality assurance, but these conditions were routinely 

reported as damage. More recent investigations strongly suggest that these weld root flaws are 

not earthquake damage, but defects from the original construction. Only one third of the total 

reports prepared under the Los Angeles City ordinance indicated damage other than weld root 

defects. Of the buildings with reported damage other than weld root defects, two-thirds had less 

than 10% of their connections fractured. Only 11% of all the buildings included in the ordinance 

had more than 10% of their connections damaged, while relatively few buildings (13% of the 

total) accounted for 90% of all damage other than defects at the weld roots. 

The distribution of damage in these buildings points to some important potential findings. 

The concentration of severe damage in a relatively small percentage of the total buildings 

inspected would seem to indicate that in order to sustain severe damage, a steel moment-frame 

building must either experience very strong response to the earthquake ground motion, or, as a 

result of design configuration or construction quality, or both, be particularly susceptible to 

damage. It would seem that most steel moment-frame buildings are not particularly susceptible 

to severe damage under ground shaking of Modified Mercalli Intensity VII or less. 

Although initial reports following the 1994 Northridge earthquake indicated that more than 

100 buildings had sustained severe damage, in many cases this reported damage was limited to 

discontinuities and defects at the root of the complete joint penetration (CJP) welds between the 

beam bottom flange and the column flange. As previously noted, there is strong evidence to 

suggest that most such conditions are not damage at all, but rather, pre-existing construction 

defects that were not detected during the original construction quality assurance program. 

Subsequent research in other buildings and cities suggests that the presence of such defects is 

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widespread and generally present in the population of welded steel moment-frame (WSMF) 

buildings constructed in the United States prior to the Northridge earthquake. 

Notwithstanding the above comments, a number of buildings did experience brittle fracture 

damage in their beam-column connections. The amount of damage sustained by buildings was 

generally related to the severity of ground shaking experienced at the building site as well as the 

severity of response of the structure to the ground shaking, although this second factor was not 

necessarily measured during the earthquake. However, the presence of construction defects in 

the welded joints was also a significant factor in the initiation of fracture damage. Joints with 

severe defects at the weld roots were more susceptible to fracture initiation than joints without 

such defects. Since the distribution of joints with defects in an existing structure is somewhat 

random, this tends to minimize the effectiveness of structural analysis in predicting the exact 

locations where damage is likely to occur under ground shaking. However, probabilistic 

methods based on structural analysis can be successful in indicating the general likelihood of 

damage, given certain levels of ground shaking. Therefore, the evaluation and design criteria 

contained in these Recommended Criteria are based on such probabilistic approaches. 

Commentary: Detailed information on the types of damage discovered in various 

WSMF buildings following past earthquakes may be found in a companion report, 

FEMA-355E - State of the Art Report on Past Performance of Steel MomentFrame Buildings in Earthquakes. 

2.2.4 Damage to Welded Steel Moment-Frame (WSMF) Construction in Other 

Earthquakes 

Following the discovery of unanticipated damage to WSMF construction in the 1994 

Northridge earthquake, engineers and building officials became concerned that similar, but as yet 

undetected damage, had occurred in WSMF buildings that had been affected by other 

earthquakes, such as the 1989 Loma Prieta earthquake in the San Francisco Bay Area. A 

concerted effort was undertaken by this project to determine the amount and extent of earthquake 

damage resulting from this and other recent earthquakes. Specifically, available WSMF damage 

information was gathered from the 1989 Loma Prieta, 1992 Landers, and 1992 Big Bear events. 

Unfortunately, since no mandatory inspection programs of WSMF buildings were enacted 

following these other earthquakes, the available data is not complete. It was, however, possible 

to confirm that six buildings in the San Francisco Bay Area sustained connection fractures in the 

Loma Prieta earthquake and one building in Big Bear, California sustained connection fractures 

as a result of the 1992 events. This confirms that the damage experienced in the 1994 Northridge 

earthquake was not a result either of unique ground shaking characteristics produced by that 

earthquake or of design and construction practices unique to the Los Angeles region. Further 

details of these investigations may be found in FEMA-355E. 

One year to the day following the Northridge earthquake, on January 17, 1995, a magnitude 

6.9 earthquake occurred near Kobe, Japan. Kobe is a large city with a population of about 1.5 

million and had many WSMF structures in its building stock. These structures ranged from 

relatively small and low-rise buildings constructed in the 1950s and 1960s to modern high-rise 

structures constructed within the preceding 10 years. Design and construction practice in Japan 

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is significantly different from common practice in the United States. Many of the smaller 

Japanese steel moment-frame (WSMF) structures employ cold-formed, tubular steel columns, 

with the beams, rather than columns, running continuously through the moment-resisting 

connections. In a detailed study of the damage sustained by 630 modern steel buildings in the 

heavily shaken area, the Building Research Institute of Japan determined that approximately one 

third experienced no significant damage, one third relatively minor damage, and the remaining 

third severe damage, including partial or total collapse of approximately half of the buildings in 

this remaining third (FEMA-355E). Just as in the United States, the Japanese believed that this 

damage was serious enough to warrant investment in a large program of research and 

development to determine the cause of the poor performance of WSMF buildings and to develop 

new techniques for design and construction of more reliable WSMF buildings. 

2.2.5 Post-Northridge Earthquake Construction Practice 

Investigation of the damage that occurred in the 1994 Northridge earthquake revealed a 

number of factors believed to have contributed to the poor performance of WSMF structures. 

These included the following: 

• It was common practice to use large framing members even in relatively small buildings. 

Initial testing of WSMF connections, conducted in the 1960s and 1970s, utilized assemblies 

that employed small-sized elements, typically W18 beams and light W12 and W14 column 

sections. Typical buildings damaged by the Northridge earthquake employed W30 or larger 

beams connected to heavy W14 columns. It appears that size plays a significant role in the 

behavior of WSMF connections and that details that behave well for connections of small 

sections do not necessarily behave as well for larger sections. 

• Typical detailing practice prior to the Northridge earthquake relied on the development of 

large inelastic behavior within the beam-column connections. This was the case even though 

one of the basic rules of detailing structures for superior seismic performance is to design 

connections of elements such that the connection is stronger than the elements themselves, so 

that any inelastic behavior occurs within the element and not the connection. There are 

several reasons for this rule. The strength and ductility of any connection is highly dependent 

on the quality of the workmanship employed. Connections, being relatively limited in size, 

must undergo extreme local yielding if they are to provide significant global ductility. The 

basic fabrication process for connections, employing cutting, welding, and bolting, tends to 

induce a complex series of effects on both the residual stress state and metallurgy of the 

connected parts that is often difficult to predict. Despite these common axioms of 

earthquake-resistant design the connections were called on for large inelastic behavior. 

• Welding procedures commonly employed in the erection of WSMF buildings resulted in 

deposition of low-notch-toughness weld metal in the critical beam-flange-to-column-flange 

joints. This weld metal is subject to the initiation and development of unstable brittle 

fractures when subjected to high stress and strain demands and used in situations with 

significant geometric stress risers, or notches. 

• Welding practice in many of the damaged structures was found to be sub-standard, despite 

the fact that quality assurance measures had been specified in the construction documents and 

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that construction inspectors had signed documents indicating that mandatory inspections had 

been performed. Damaged welds commonly displayed inadequate fusion at the root of the 

welds as well as substantial slag inclusions and porosity. These defects resulted in ready 

crack initiators that enabled brittle fractures to initiate in the low-toughness weld metals. 

• Detailing practice for welded joints inherently resulted in the presence of fracture-initiators. 

This includes failure to remove weld backing and runoff tabs from completed joints. These 

joint accessories often contain or obscure the presence of substandard welds. In addition, 

they introduce geometric conditions that are notch-like and can serve as fracture initiators. 

• The presence of low-notch-toughness metal in the fillet region of some structural shapes can 

contribute to early fractures. The metallurgy of the material in the fillet or “k-area” region of 

a rolled shape often has lower notch toughness properties than material in other locations of 

the section due to a number of shape production factors including a relatively prolonged 

cooling period for this area, as well as significant cold working during shape straightening. 

While not normally a problem, the combined presence of weld access holes through this 

region at the beam-column connection and large induced stresses from buckling and yielding 

of the beam flanges under inelastic frame action can result in initiation of fractures in this 

region. These problems are made more severe by improperly cut weld access holes, which 

can result in sharp notches and crack initiation points. This was not a common problem in 

the Northridge earthquake because most connections that experienced damage did so because 

of other, more significant vulnerabilities. However, some of the damage that occurred to 

Japanese structures in the Kobe earthquake was apparently the result of these problems. 

• In the 1980s, some engineers came to believe that shear yielding of the panel zones in a 

beam-column connection, as opposed to flexural hinging of the beam, was a more benign and 

desirable way to accommodate frame inelastic behavior. In response to this, in the mid-1980s 

the building code was modified to include provisions that allowed the design of frames with 

weak panel zones. Contrary to the belief that panel-zone yielding is beneficial and desirable, 

excessive yielding actually produces large secondary stresses at the beam-flange-to-columnflange joint, which can exacerbate the initiation of fractures. 

• The yield strength of structural shape material had become highly variable. In the 1980s and 

1990s, the steel production industry in the United States underwent a major realignment with 

new mills coming on-line and replacing older mills. Although there had always been 

significant variation in the mechanical properties of structural steel material, the introduction 

of material produced by these newer mills resulted in significant additional variation. The 

newer mills used scrap-based steel production, which tends to produce higher-strength 

material than did the older mills. In fact, much of the A36 material produced by these newer 

mills also met the strength requirements for the higher strength A572, Grade 50 specification. 

Many designers had traditionally specified A572 material for columns and A36 material for 

beams, in order to obtain structures economically with weak beams and strong columns. The 

introduction of higher strength A36 material into the market effectively negated the intent of 

this specification practice and often resulted in frame assemblies in which the beams were 

stronger than the columns or panel zones were weaker than intended, relative to the beam 

strength. These combined effects resulted in greater strength demands on welded joints. 

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• The typical steel moment-frame beam-column connection inherently incorporated a number 

of stress concentrations. Although design calculations of connection capacity assume that 

stresses are uniformly distributed across beam flanges and that flexural stresses are carried 

primarily by the flanges while shear stresses are carried primarily by the web, in reality, the 

flange also carries significant local bending and shear stress and stresses are not uniformly 

distributed within the flange elements. The result of this is that large stress and strain 

demands occur at various locations, including the center of the weld root of the welded beamflange-to-column-flange joint. This exacerbates the tendency of the weld defects, which are 

common in this region, to initiate brittle fractures in the low-notch-toughness metal. This 

effect is further exacerbated by the fact that the material at the center of the beam-flange-tocolumn-flange joint is under high tri-axial restraint. Under these conditions the material 

cannot yield, but rather will respond to stress in an elastic manner until the ultimate tensile 

strength is exceeded, at which time it initiates fracture. This problem is most severe when 

heavy sections are used, as the thicker material provides greater restraint. 

Following the discovery of the susceptibility of typical pre-Northridge connections to fracture 

damage, an emergency change to the Uniform Building Code was adopted by the International 

Conference of Building Officials, removing the prequalified status of the typical bolted-web, 

welded-flange moment connection previously prescribed by the code and substituting in its place 

requirements that each connection design be qualified by a program of prototype laboratory 

testing. In 1994, the University of Texas at Austin engaged in a limited program of connection 

testing, using funding provided by the American Institute of Steel Construction and a private 

institution. That testing indicated that connections reinforced with cover plates to encourage the 

formation of plastic behavior within the span of the beam, away from the face of the columns, 

could provide acceptable behavior. This detail is illustrated in Figure 2-3. During the period 

1994-1996 this became the most commonly specified connection type. 

In the earliest connections of this type, welding was performed with electrodes that deposited 

material without rated notch toughness and with a wide variety of cover plate configurations. In 

August, 1995, FEMA-267 was published, providing a standardized methodology for design of 

these connections, and the design and fabrication of these connections became more consistent. 

FEMA-267 required the use of weld filler metals with rated notch toughness, and also included 

information on other types of connections that were believed capable of providing acceptable 

performance, including haunched connections, reduced-beam-section connections, vertical rib 

plate connections, side plate connections and slotted web connections. The recommendations 

contained in FEMA-267 were based on preliminary research and were of an interim nature. 

While it is expected that frames constructed with connections designed using the FEMA-267 

guidelines are more resistant to connection fractures than earlier frames, it should not be assumed 

that they are completely free of potential for such damage. 

Subsequent to the publication of FEMA-267, numerous other connection types have been 

developed and tested. For the upgrade of existing buildings, solutions utilizing connection 

modifications are discussed in Chapter 5 of these Recommended Criteria and supporting 

information is presented in Chapter 6, Connection Qualification. 

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Figure 2-3 Typical Cover Plate Connection 

2.3 Typical Pre-Northridge Connection Damage 

Following the 1994 Northridge earthquake, damage to elements of welded steel moment 

frames (WSMF) was generally categorized according to a system published in FEMA-267. 

Under this system, damage is categorized as belonging to the weld (W), girder (G), column (C), 

panel zone (P), or shear tab (S) categories. Damage at a connection may be confined to one 

category or may include multiple types. The damaged WSMF may also exhibit global effects, 

such as permanent interstory drifts. The components of a typical pre-Northridge connection are 

shown in Figure 2-4. 

Observation of damage sustained by buildings in the Northridge earthquake indicates that in 

many cases brittle fractures initiated within the connections at very low levels of plastic demand, 

and in some cases, while the structures remained elastic. Typically, but not always, fractures 

initiated at the complete joint penetration (CJP) weld between the beam bottom flange and 

column flange as shown in Figure 1-2. Once initiated, these fractures progressed along a number 

of different paths, depending on the individual joint conditions. 

In some cases, the fracture progressed completely through the thickness of the weld, and if 

fire protective finishes were removed, the fracture was evident as a crack through exposed faces 

of the weld, or the metal just behind the weld (Figure 1-3a). Other fracture patterns also 

developed. In some cases, the fracture developed into a through-thickness failure of the column 

flange material behind the CJP weld (Figure 1-3b). In these cases, a portion of the column flange 

remained bonded to the beam flange, but pulled free from the remainder of the column. This 

fracture pattern has sometimes been termed a “divot” or “nugget” failure. 

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Panel 

Zone 

Beam 

Column 

Figure 2-4 Components of Moment Connection 

A number of fractures progressed completely through the column flange, along a nearhorizontal plane that aligns approximately with the beam lower flange (Figure 1-4a). In some 

cases, these fractures extended into the column web and progressed across the panel zone Figure 

(1-4b). Investigators have reported some instances where columns fractured entirely across the 

section. 

Once such fractures have occurred, the beam-column connection has experienced a 

significant loss of flexural rigidity and strength. Residual flexural strength and rigidity must be 

developed through a couple consisting of forces transmitted through the remaining flange 

connection and the web bolts. However, in providing this residual strength and stiffness, the 

beam shear connections can themselves be subject to failures, consisting of fracturing of the 

welds of the shear plate to the column, fracturing of supplemental welds to the beam web or 

fracturing through the weak section of shear plate aligning with the bolt holes (Figure 1-5). 

Despite the obvious local strength impairment resulting from these fractures, many damaged 

buildings did not display overt signs of structural damage, such as permanent drifts, or extreme 

damage to architectural elements. The following sections detail typical damage types, using the 

system for categorizing damage recommended in FEMA-352 – Recommended Postearthquake 

Evaluation and Repair Criteria for Welded Steel Moment-Frame Buildings for postearthquake 

damage assessment. 

2.3.1 Girder Damage 

Girder damage may consist of yielding, buckling or fracturing of the flanges of girders at or 

near the girder-column connection. Eight separate types are defined in Table 2-1. Figure 2-5 

illustrates these various types of damage. 

Minor yielding of girder flanges (type G2) is the least significant type of girder damage. It is 

often difficult to detect and may be exhibited only by local flaking of mill scale and the formation 

of characteristic visible lines in the material, running across the flange. If a finish, or 

fireproofing has been removed by scraping, the detection of this type of damage is difficult. 

Backing 

Shear Tab 

Supplemental weld 

Continuity Plates 

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Girder flange yielding, without local buckling or fracture, results in negligible degradation of 

structural strength. 

Table 2-1 Types of Girder Damage 

Type Description 

G1 Buckled flange (top or bottom) 

G2 Yielded flange (top or bottom) 

G3 Flange fracture in heat affected zone (HAZ) (top or 

bottom) 

G4 Flange fracture outside heat affected zone (HAZ) 

(top or bottom) 

G5 Flange fracture top and bottom (not used) 

G6 Yielding or buckling of web 

G7 Fracture of web 

G8 Lateral torsional buckling of section 

G1 

G3 G2

G4 

G7 

G6 

G8

Figure 2-5 Types of Girder Damage 

Girder flange buckling (type G1) can result in a significant loss of girder plastic strength, 

particularly when accompanied by girder web buckling (type G-6). For compact sections, this 

strength loss occurs gradually, and increases with the number of inelastic cycles and the extent of 

the inelastic excursion. Following the initial onset of buckling, additional buckling will often 

occur at lower load levels and result in further reductions in strength, from the levels of previous 

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cycles. The localized secondary stresses which occur in the girder flanges due to the buckling 

can result in initiation of flange fracture damage (G4) if the frame is subjected to a large number 

of cycles. Such fractures typically progress slowly, over repeated cycles, and grow in a ductile 

manner. Once this type of damage initiates, the girder flange will begin to lose tensile capacity 

under continued or reversed loading, although it may retain some capacity in compression. 

In structures with low-toughness welds, girder flange cracking within the heat-affected zone 

(type G3) can occur as an extension of brittle fractures that initiate in the weld root. This is 

particularly likely to occur at connections in which improper welding procedures were followed, 

resulting in a brittle heat-affected zone. However, these fractures can also occur in connections 

with tough welded joints (made following appropriate procedures), as a result of low-cycle 

fatigue, exacerbated by the high stress concentrations that occur at the toe of the weld access 

hole, in unreinforced beam-column connections. Like the visually similar type G4 damage, 

which can also result from low-cycle fatigue conditions at the toe of the weld access hole, it 

results in a complete loss of flange tensile capacity, and consequently significant reduction in the 

contribution to frame lateral strength and stiffness from the connection. 

In the 1994 Northridge earthquake, girder damage was most commonly detected at the 

bottom flanges, although some instances of top flange failure were also reported. There are 

several reasons for this. First the composite action induced by the presence of a floor slab at the 

girder top flange tends to shift the neutral axis of the beam towards the top flange. This results in 

larger tensile deformation demands on the bottom flange than on the top. In addition, the 

presence of the slab tends to reduce the chance of local buckling of the top flange. The bottom 

flange, however, being less restrained can experience buckling relatively easily. 

There are a number of other factors that could lead to a greater incidence of bottom flange 

fractures. The location of the weld root and backing are among the most important of these. At 

the bottom flange joint, the backing is located at the extreme tension fiber, while at the top flange 

it is located at a point of lesser stress and strain demand for three reasons: (1) it is located on the 

inside face of the flange, (2) the local bending introduced in the flanges as a result of panel zone 

shear deformations, and (3) because of the presence of the floor slab. Therefore, any notch 

effects created by root defects and backing are more severe at the bottom flange. Another 

important factor is that welders can typically make the complete joint penetration groove weld at 

the girder top flange without obstruction, while the bottom flange weld must be made with the 

restriction induced by the girder web. Also the welder typically has better access to the top 

flange joint. Thus, top flange welds tend to be of higher quality, and have fewer stress risers, 

which can initiate fracture. Finally, studies have shown that inspection of the top flange weld is 

more likely to detect defects accurately than inspection at the bottom flange, contributing to the 

better quality likely to occur in top flange welds. 

2.3.2 Column Flange Damage 

Seven types of column flange damage are defined in Table 2-2 and illustrated in Figure 2-6. 

Column damage typically results in degradation of a structure’s gravity-load-carrying strength as 

well as lateral-load resistance. 

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Table 2-2 Types of Column Damage 

Type Description 

C1 Incipient flange crack 

C2 Flange tear-out or divot 

C3 Full or partial flange crack outside heataffected zone 

C4 Full or partial flange crack in heat-affected 

zone 

C5 Lamellar flange tearing 

C6 Buckled flange 

C7 Column splice failure 

C1 

C2 

C3 

C4 

C5 

C6 

C7 

Figure 2-6 Types of Column Damage 

Column flange damage includes types C1 through C7. Type C1 damage consists of a small 

crack within the column flange thickness, typically at the location of the adjoining girder flange. 

C1 damage does not go through the thickness of the column flange and can be detected only by 

nondestructive testing. Type C2 damage is an extension of type C1, in which a curved failure 

surface extends from an initiation point, usually at the root of the girder-to-column-flange weld, 

and extends longitudinally along the column flange. In some cases this curved failure surface 

may emerge on the same face of the column flange as the one where it initiated. When this 

occurs, a characteristic nugget or divot can be withdrawn from the flange. Types C3 and C4 

fractures extend through the thickness of the column flange and may extend into the panel zone. 

Type C5 damage is characterized by a step-shaped failure surface within the thickness of the 

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column flange and aligned parallel to it. This damage is often detectable only with the use of 

nondestructive testing. 

Type C1 damage does not result in an immediate large strength loss in the column; however, 

such small fractures can easily progress into more serious types of damage if subjected to 

additional large tensile loading by aftershocks or future earthquakes. Type C2 damage results in 

both a loss of effective attachment of the girder flange to the column for tensile demands and a 

significant reduction in available column flange area for resistance of axial and flexural demands. 

Type C3 and C4 damage result in a loss of column flange tensile capacity and, under additional 

loading, can progress into other types of damage. 

Type C5, lamellar tearing damage, may occur as a result of non-metallic inclusions within the 

column flange, particularly in older steels, when, prior to rolling, segregation of alloy inclusions 

was not controlled as well as in modern steels. The potential for this type of fracture under 

conditions of high restraint and large through-thickness tensile demands has been known for a 

number of years and has sometimes been identified as a contributing mechanism for type C2 

column flange through-thickness failures. No lamellar tearing failures were identified after the 

Northridge earthquake. 

Type C6 damage consists of local buckling of the column flange, adjacent to the beamcolumn connection. While such damage was not actually observed in buildings following the 

1994 Northridge earthquake, it can be anticipated at locations where plastic hinges form in the 

columns. Buckling of beam flanges has been observed in the laboratory at interstory drift 

demands in excess of 0.02 radians. Column sections are usually more compact than beams and 

therefore are less prone to local buckling. Type C6 damage may occur, however, in buildings 

with strong-beam-weak-column systems and at the bases of columns in any building when large 

interstory drifts have occurred. 

Type C7 damage, fracturing of welded column splices, also was not observed following the 

Northridge earthquake. However, the partial penetration groove welds commonly used in these 

splices are susceptible to fracture when subjected to large tensile loads. Large tensile loads can 

occur on a column splice as a result of global overturning effects, or as a result of large flexural 

demands in the column. 

2.3.3 Weld Damage, Defects, and Discontinuities 

Three types of weld damage are defined in Table 2-3 and illustrated in Figure 2-7. All apply 

to the complete joint penetration welds between the girder flanges and the column flanges. 

Type W2 fractures extend completely through the thickness of the weld metal and can be 

detected by either Magnetic Particle Testing (MT) or Visual Inspection (VI) techniques. Type 

W3 and W4 fractures occur at the zone of fusion between the weld filler metal and base material 

of the girder and column flanges, respectively. All three types of damage result in a loss of 

tensile capacity of the girder-flange-to-column-flange joint. 

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Table 2-3 Types of Weld Damage, Defects and Discontinuities 

Type Description 

W2 Crack through weld metal thickness 

W3 Fracture at column interface 

W4 Fracture at girder flange interface 

W2 

W3 W4 

Figure 2-7 Types of Weld Damage 

In addition to the W2, W3, and W4 types of damage indicated in Table 2-3 and Figure 2-7, 

the damage classification system presented in FEMA-267 included conditions at the root of the 

complete joint penetration weld that did not propagate through the weld nor into the surrounding 

base metal, and could be detected only by removal of the weld backing or through the use of 

nondestructive testing. These conditions were termed types W1a, W1b, and W5. 

As defined in FEMA-267, type W5 consisted of small discontinuities at the root of the weld, 

which, if discovered as part of a construction quality control program for new construction would 

not be rejectable under the AWS D1.1 provisions. FEMA-267 recognized that W5 conditions 

were likely to be the result of acceptable flaws introduced during the initial building construction, 

but included this classification so that it could be reported in the event that it was detected in the 

course of the ultrasonic testing that FEMA-267 required. There was no requirement to repair 

such conditions. 

Type W1a and W1b conditions, as contained in FEMA-267 consisted of discontinuities, 

defects and cracks at the root of the weld that would be rejectable under the AWS D1.1 

provisions. W1a and W1b were distinguished from each other only by the size of the condition. 

Neither condition could be detected by visual inspection unless weld backing was removed, 

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which, in the case of W1a conditions, would also result in removal of the original flaw or defect. 

At the time FEMA-267 was published, there was considerable controversy as to whether or not 

the various types of W1 conditions were actually damage or just previously undetected flaws 

introduced during the original construction. Research conducted since publication of FEMA-267 

strongly supports the position that most, if not all, W1 damage consists of pre-existing defects, 

rather than earthquake damage. 

2.3.4 Shear Tab Damage 

Six types of damage to girder-web-to-column-flange shear tabs are defined in Table 2-4 and 

illustrated in Figure 2-8. Severe damage to shear tabs is often an indication that other damage 

has occurred to the connection, i.e., to the column, girder, panel zone, or weld. 

Table 2-4 Types of Shear Tab Damage 

Type Description 

S1 Partial crack at weld to column 

S2 Fracture of supplemental weld 

S3 Fracture through tab at bolts or severe 

distortion 

S4 Yielding or buckling of tab 

S5 Loose, damaged or missing bolts 

S6 Full length fracture of weld to column 

Shear tab damage should always be considered significant, as failure of a shear tab 

connection can lead to loss of gravity-load-carrying capacity for the girder, and potentially partial 

collapse of the supported floor. Severe shear tab damage typically does not occur unless other 

significant damage has occurred at the connection. If the girder flange joints and adjacent base 

metal are sound, they prevent significant differential rotations from occurring between the 

column and girder. This protects the shear tab from damage, unless excessively large shear 

demands are experienced. If these excessive shear demands do occur, than failure of the shear 

tab is likely to trigger distress in the welded joints of the girder flanges. 

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S1 

S3 S2

S4 

S6 S5 

Figure 2-8 Types of Shear Tab Damage 

2.3.5 Panel Zone Damage 

Nine types of damage to the column web panel zone and adjacent elements are defined in 

Table 2-5 and illustrated in Figure 2-9. This class of damage can be among the most difficult to 

detect since elements of the panel zone may be obscured by beams framing into the weak axis of 

the column. 

Fractures in the welds of continuity plates to columns (type P2), or damage consisting of 

fracturing, yielding, or buckling of the continuity plates themselves (type P1) may be of relatively 

little consequence to the structure, so long as the fracture does not extend into the column 

material itself. Fracture of doubler plate welds (type P4) is more significant in that this results in 

a loss of effectiveness of the doubler plate and the fractures may propagate into the column 

material. 

Although shear yielding of the panel zone (type P3) is not by itself undesirable, under large 

deformations such shear yielding can result in kinking of the column flanges and can induce large 

secondary stresses in the girder-flange-to-column-flange connection. In testing conducted at the 

University of California at Berkeley, excessive deformation of the column panel zone was 

identified as a contributing cause to the initiation of type W2 fractures at the top girder flange. It 

is reasonable to expect that such damage could also be initiated in real buildings, under certain 

circumstances. 

Fractures extending into the column web panel zone (types P5, P6 and P7) have the potential, 

under additional loading, to grow and become type P9, a complete disconnection of the upper 

half of the column within the panel zone from the lower half, and are therefore potentially as 

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severe as column splice failures. When such damage has occurred, the column has lost all tensile 

capacity and its ability to transfer shear is severely limited. Such damage results in a total loss of 

reliable seismic capacity. It appears that such damage is most likely to occur in connections that 

are subject to column tensile loads, or in connections in which beam yield strength exceeds the 

yield strength of the column material. 

Table 2-5 Types of Panel Zone Damage 

Type Description 

P1 Fracture, buckle or yield of continuity plate 

P2 Fracture in continuity plate welds 

P3 Yielding or ductile deformation of web 

P4 Fracture of doubler plate welds 

P5 Partial depth fracture in doubler plate 

P6 Partial depth fracture in web 

P7 Full or near full depth fracture in web or doubler 

P8 Web buckling 

P9 Severed column 

P2 

P7 

P9 

P4

P8 

P3 P5, P6 P1

Figure 2-9 Types of Panel Zone Damage 

Panel-zone web buckling (type P8) may result in rapid loss of shear stiffness of the panel 

zone with potential total loss of reliable seismic capacity. Such buckling is unlikely to occur in 

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connections which are stiffened by the presence of a vertical shear tab for support of a beam 

framing into the column’s minor axis. 

2.3.6 Other Damage 

In addition to the types of damage discussed in the previous sections, other types of structural 

damage may also be found in steel moment-frame buildings. Other framing elements which may 

experience damage include: (1) column base plates, beams, columns, and their connections that 

were not intended in the original design to participate in lateral force resistance, and (2) floor and 

roof diaphragms. In addition, large permanent interstory drifts may develop in the structures. 

Based on observations of structures affected by the 1994 Northridge earthquake, such damage is 

unlikely unless extensive damage has also occurred to the lateral-force-resisting system. 

2.4 Evaluation Procedures 

This document provides recommendations for performing several types of evaluation of the 

probable performance of existing steel moment-frame buildings in future earthquakes, as 

outlined below: 

• Performance Evaluation. The purpose of performance evaluation is to permit estimation of 

a level of confidence that a structure will be able to achieve a desired performance objective 

(i.e., have less than a given probability of experiencing damage in excess of one or more 

defined limit states). In these Recommended Criteria, building damage is characterized in 

terms of two performance levels. Section 3.2.2 provides definitions of these performance 

levels. Once a performance objective for a building has been selected, a performance 

evaluation can be performed in accordance with Section 3.3 to determine a level of 

confidence with regard to the structure’s ability to meet this performance objective. The 

level of confidence that can be attained with regard to the ability of a building to meet a 

desired performance objective is dependent on the amount of information that is available 

with regard to the building’s configuration and construction, and the rigor of the analytical 

methods used in the evaluation. The performance evaluation procedures contained in Section 

3.3 include simple methods for the quantification of uncertainty and confidence with regard 

to performance prediction of regular, well-behaved structures. More detailed methods, that 

permit more certain evaluation of performance capability, and which must be used for 

evaluation of irregular buildings are contained in Appendix A. Procedures and information 

regarding material properties and condition assessments to be utilized in support of the 

performance evaluation are presented in Section 2.5. 

Commentary: In recent years, a series of standardized building performance 

evaluation methodologies, including ATC-14, FEMA-154, FEMA-178 and most 

recently FEMA-310, have been developed. These methodologies were developed 

to provide the engineering community with consistent yet economical methods of 

determining the probable performance of different types of buildings when 

subjected to specific earthquake ground shaking levels. Evaluations performed in 

accordance with these methodologies generally consist of responding to a series 

of evaluation statements, intended to identify the presence of certain common 

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vulnerabilities, such as soft stories, weak stories, and discontinuous lateral-forceresisting systems that have been frequently observed to result in poor building 

performance in the past. These methodologies also commonly employ a series of 

analytical evaluations that include approximate evaluations of building strength 

and stiffness. 

While these methodologies provide good screening criteria to identify those 

buildings that have obvious vulnerabilities, and also serve to identify those 

buildings that have outstanding seismic performance characteristics, the 

approximate analytical procedures employed in these methods inherently 

incorporate so much uncertainty as to make them relatively ineffective for 

quantifying building performance. 

Nevertheless, it is recommended that FEMA-310 be performed as a first step 

in the analytical evaluation of a building’s probable seismic performance. Such 

an evaluation will provide the engineer with a basic understanding of potential 

critical flaws in the building configuration and provide a basis for a more 

detailed analytical evaluation of the building’s performance, under the 

procedures of these Recommended Criteria. 

• Loss Evaluation. The purpose of a loss evaluation is to determine the probable repair costs 

for a structure (or class of structures), if it is subjected to an earthquake hazard of defined 

intensity. In most loss-estimation methodologies, repair costs are expressed as a percentage 

of the building replacement cost. Loss-estimation evaluations sometimes include estimates 

of potential interruption of building occupancy as well as repair cost. Two approaches to loss 

estimation are provided herein: a rapid loss-estimation methodology and a detailed lossestimation method. Rapid loss estimation, described in Chapter 4, can be quickly performed 

using basic data on the building’s construction characteristics and specification of the 

intensity of ground shaking for which the loss evaluation is being performed. Detailed loss 

estimation requires an analytical evaluation of the building and estimation of the ground 

shaking response accelerations at which different damage states are likely to be exceeded. 

Appendix B provides information on detailed loss-estimation methods that are compatible 

with HAZUS, FEMA’s nationally applicable earthquake-loss-estimation model. 

Commentary: The rapid loss evaluation methodology is an approach similar to 

that taken in ATC-13 (ATC, 1985), in which the probability of experiencing a 

certain loss is related to the intensity of ground shaking experienced at the site, 

measured by the Modified Mercalli Intensity (MMI). Such methodologies were 

originally developed to estimate the probable distribution of losses for broad 

classes or populations of buildings. These methodologies are generally based on 

either actuarial statistics of the actual losses experienced by populations of 

buildings in past earthquakes, or on statistics related to expert opinion on the 

probable performance of actual buildings, or both. The methods have no direct 

way to account for individual building structural performance characteristics 

such as strength, stiffness, redundancy, or regularity, and as a result, inherently 

incorporate a great deal of uncertainty when applied to estimation of the loss for 

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a specific building structure. However, in recent years, the application of these 

methodologies to single building loss estimation, though technically incorrect, 

has become common. This application is not recommended. 

The detailed loss-estimation methodology presented in Appendix B provides 

for the direct consideration of structural characteristics, important to building 

performance, in the loss-evaluation process. In this methodology, structural 

analyses of the building structure are performed to characterize the probable 

response of the building to ground motion. Statistical data are then used to relate 

building response to damage and loss, at defined levels of uncertainty. The 

detailed loss-estimation methodology is recommended for applications in which it 

is desired to estimate the probable losses for a single building, as opposed to 

populations of buildings. It is particularly recommended as a design verification 

methodology for those cases when it is desired to upgrade a building to protect 

against future economic loss. 

2.5 Material Properties and Condition Assessments 

In order to perform a meaningful evaluation of either type, it is necessary to understand the 

structure’s basic configuration, its condition, and certain basic material properties. The extent of 

the necessary knowledge depends on the type of evaluation and the level of certainty desired for 

the conclusions drawn from the evaluation. Original construction documents, including the 

drawings and specifications will provide sufficient data for the evaluation of most steel momentframe buildings, so long as the building was actually constructed in accordance with these 

documents. As a minimum, the evaluation should include at least one visit to the building site to 

determine its overall condition and to confirm that available record documents are reasonably 

representative of the actual construction. If no construction documents are available, then 

extensive field surveys may be required to define the structure’s configuration, including the 

locations of frames, the sizes of framing elements and connection details, as well as the materials 

of construction. 

2.5.1 Material Properties 

The primary material properties required to perform analytical evaluations of a steel momentframe building include the following: 

• yield strength, ultimate tensile strength and modulus of elasticity of steel for the columns in 

the moment frames, 

• yield strength, ultimate tensile strength and modulus of elasticity of steel for the beams in the 

moment frames, 

• ultimate tensile strength and notch toughness of the weld metal in the moment-resisting 

connections, and 

• yield and ultimate tensile strength of bolts in the moment-resisting connections. 

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Although structural steel is an engineered material, there can be significant variability in the 

properties of the steel in a building, even if all of the members and connection elements conform 

to the same specifications and grades of material. Exhaustive programs of material testing to 

quantify the physical and chemical properties of individual beams, columns, bolts, and welds are 

not justified and should typically not be performed. It is only necessary to characterize the 

properties of material in a structure on the basis of the likely statistical distributions of the 

properties noted above, with mean values and coefficients of variation. Knowledge of the 

material specification and grade that a structural element conforms to, and its approximate age 

will be sufficient to define these properties for nearly all evaluations. For rapid loss-estimation 

evaluations, it will not be necessary to determine material properties. 

In general, analytical evaluations of global building behavior are performed using expected or 

mean values of the material properties (based on the likely distribution of these properties) for 

the different grades of material present in the structure. Expected values are denoted in these 

procedures with the subscript “e”. Thus, the expected yield and ultimate tensile strength of steel 

are denoted, respectively, Fye and Fue. Some calculations of individual connection capacities are 

performed using lower-bound values of strength. Where lower-bound strength values are 

required, the yield and tensile strength are denoted as Fy and Fu, respectively. Lower-bound 

strengths are defined as the mean minus two standard deviations, based on statistical data for the 

particular specification and grade. 

If original construction documents, including drawings and specifications, are available, and 

indicate in an unambiguous manner the materials of construction to be employed, it will typically 

not be necessary to perform materials testing in a steel moment-frame building. When material 

properties are not clearly indicated on the drawings and specifications, or the drawings and 

specifications are not available, the material grades indicated in Table 2-6 may be presumed. 

Alternatively, a limited program of material sample removal and testing may be conducted to 

confirm the likely grades of these materials. 

If sampling is performed, it should take place in regions of reduced stress, such as flange tips 

at ends of simply supported beams, flange edges in the mid-span region of members of momentresisting frames, and external plate edges, to minimize the effects of the reduced area. If a bolt is 

removed for testing, a comparable bolt should be reinstalled in its place. If coupons are removed 

from beams or columns, the material should either be replaced with the addition of reinforcing 

plate, or the area of removal should be dressed to provide smooth contours of the cutout area, 

without square corners or notches. Removal of a welded connection sample must be followed by 

repair of the connection. When sampling is performed to confirm the grades of material present 

in a structure, mechanical properties should be determined in the laboratory using industry 

standard procedures in accordance with ASTM A-370. 

For the purpose of analytical evaluation of steel moment-frame buildings, the expected and 

lower bound strength of structural materials shall be taken from Table 2-7, based on the age, 

material specification, and grade of material. 

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Commentary: In general, great accuracy in the determination of the material 

properties of structural steel elements in steel moment-frame (WSMF) buildings is 

neither justified nor necessary, in order to perform reasonably reliable 

evaluations of building performance. The two most important parameters are the 

yield strengths of the beams and columns and the notch toughness of the weld 

metal. 

Weld Filler Metal 

Welding was first introduced into the building construction industry in the 

early 1950s. Prior to that time, most structural connection was made either by 

riveting or bolting. Early structural welding typically used the shielded metal arc 

welding (SMAW) process and “stick” filler metals with an ultimate tensile 

strength of 60 ksi. Although a variety of weld filler metals were available, the 

most commonly employed filler metal in the 1950s and early 1960s conformed to 

the E6012 designation. In the 1960s, as higher strength steels came on the 

market, there was a gradual shift to the E7024 weld filler metal, which was 

capable of depositing metal with a 70 ksi ultimate tensile strength. Neither of 

these filler metals had specific rating for notch toughness, although some welds 

placed with these filler metals may have considerable toughness. In the mid1960s, contractors began to switch to the semi-automatic, flux cored arc welding 

(FCAW) process, which permitted more rapid deposition of weld metal and 

therefore, more economical construction of welded structures. 

Welds in most steel moment-frame buildings constructed in the period 1964-

1994 were made with the FCAW process, employing either E70T-4 or E70T-7 

weld filler metal. This material generally has low notch toughness at service 

temperatures. Precise determination of the notch toughness of individual welds is 

not required in order to predict the probable poor performance of momentresisting connections made with these materials and the typical detailing of the 

time. However, if weld metal with significant notch toughness (40 ft-lbs at service 

temperature) has been used, even connections of the type typically constructed 

prior to the 1994 Northridge earthquake can provide limited ductility. It is rarely 

possible to determine the type of weld filler metal used in a building without 

extraction and testing of samples. Construction drawings and specifications 

typically do not specify the type of weld filler metal to be employed and even when 

they do, contractors may make substitutions for specified materials. Welding 

Procedure Specifications (WPS) for a project, if available, would define the type 

of weld filler metal employed, but these documents are rarely available for an 

existing building. Given the near universal use of the FCAW process with E70T-4 

or E70T-7 weld filler metal during the period 1964-1994, sampling of weld metal 

for buildings constructed in this period is not recommended. For buildings 

constructed prior to 1970, sampling and testing of weld filler metal may indicate 

the presence of weld with superior notch toughness, which would provide a 

higher level of confidence that the building would be capable of meeting desired 

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performance objectives. Buildings constructed prior to 1964 may conservatively 

be assumed to be constructed using weld filler metal with low toughness, or 

samples may be extracted. 

Most buildings constructed after 1996 employ weld filler metals with adequate 

notch toughness to provide ductile connection behavior. Sampling and testing of 

weld metals for buildings constructed in this period are not therefore, deemed 

necessary. During the period 1994-96, many different types of weld filler metal 

were employed in buildings. Sampling and testing of weld filler metal in 

buildings of this period may be advisable. 

When it is deemed advisable to verify the strength and notch toughness of 

weld filler metals, it is recommended that at least one weld metal sample be 

obtained and tested for each construction type (e.g., column-splice joint, or beamflange-to-column-flange joint). Samples should consist of both local base and 

weld metal, such that composite strength of the connection can be assessed. If 

ductility is required at or near the weld, the design professional may 

conservatively assume that no ductility is available in the weld, in lieu of testing. 

Beams and Columns 

The actual strength of beam and column elements in a steel moment-frame 

structure is only moderately important for the performance evaluation of such 

structures. The primary parameter used in these Recommended Criteria to 

evaluate building performance is the interstory drift induced in the building by 

earthquake ground shaking. Building drift is relatively insensitive to the actual 

yield strength of the beams and columns. However, building interstory drift can 

be sensitive to the relative yield strengths of beams and columns. In particular, 

large interstory drifts can occur in buildings with weak columns and strong 

beams, as such conditions permit the development of a single story mechanism in 

which most of the building deformation is accommodated within the single story. 

During the 1970s and 1980s, it was common practice in some regions for 

engineers to specify beams of A36 material and columns of A572, Grade 50 

material in order to develop economical designs with a strong-column-weakbeam configuration. If the properties of materials employed in a steel momentframe building are unknown, it may be conservatively assumed that the beams 

and columns are of the same specification and grade of material, in accordance 

with the default values indicated in Tables 2-6 and 2-7. However, if it can be 

determined that different grades of material were actually used for beams and 

columns, it may be possible to determine a higher level of confidence with regard 

to the ability of a building to meet desired performance objectives. In such cases, 

it may be appropriate to perform a materials sampling and testing program to 

confirm the material specifications for beams and columns. 

When it is decided to conduct a materials testing program to confirm the 

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specification and grade of material used in beams and columns, it is suggested 

that at least two strength tensile coupons should be removed from each element 

type for every four floors. If it is determined from testing that more than one 

material grade exists, additional testing should be performed until the extent of 

each grade has been established. 

Bolts 

Bolt specifications may be determined by reference to markings on the heads 

of the bolts. Where head markings are obscured, or not present, the default 

specifications indicated in Table 2-6 may be assumed. If a more accurate 

determination of bolt material is desired, a representative sample of bolts should 

be extracted from the building and subjected to laboratory testing to confirm the 

material grade. 

Table 2-6 Default Material Specifications for WSMF Buildings 

Element Type Age of Construction Default Specification 

Beams and Columns 1950-1960 ASTM A7, A373 

1961-1990 ASTM A36 

1990-1998 ASTM A572, Grade 50 

1999 and later ASTM A992 

Bolts 1950-1964 ASTM A307 

1964-1999 ASTM A325 

Weld Filler Metal 1950-1964 E6012, E7024 (1) 

1964-1994 E70T4 or E70T7 (2) 

1994-1999 See note 3 

Notes: 

1 – Prior to about 1964, field structural welding was typically performed with the Shielded Metal Arc Welding 

(SMAW) process using either E6012 or E7024 filler metal. Neither of these electrode classifications are rated 

for specific notch toughness, though some material placed using these consumables may provide as much as 

40 ft-lbs or greater notch toughness at typical service temperatures. It should be noted that due to other 

inherent characteristics of the moment resisting connection detailing prevalent prior to the 1994 Northridge 

earthquake, the presence of tough filler metal does not necessarily provide for reliable ductile connection 

behavior. 

2 – During the period 1964-1994, the Flux Cored Arc Welding (FCAW) process rapidly replaced the SMAW 

process for field welding in building structures. Weld filler metals typically employed for this application 

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conformed either to the E70T4 or E70T7 designations. Neither of these weld filler metals are rated for 

specific notch toughness, and both have similar mechanical properties. 

3 – Following the 1994 Northridge earthquake, a wide range of weld filler metals were incorporated in WSMF 

construction. Most of these filler metals had minimum ultimate tensile strengths of 70ksi and minimum rated 

notch toughness of 20 ft-lbs at –20o

F. However, due to the variability of practice, particularly in the period 

1994-1996, limited sampling of weld metal in structures in this era is recommended to confirm these 

properties. 

Table 2-7 Lower Bound and Expected Material Properties for Structural Steel Shapes of 

Various Grades 

Yield Strength (ksi) Tensile Strength (ksi) 

Material Specification Year of 

Construction 

Lower 

Bound 

Expected Lower 

Bound 

Expected 

ASTM, A7, A373 pre-1960 30 35 60 70 

ASTM, A36 Group 1 

Group 2 

Group 3 

Group 4 

Group 5 

1961-1990 41 51 60 70 

39 47 58 67 

36 46 58 68 

34 44 60 71 

39 47 68 80 

ASTM A242, A440, A441 

Group 1 

Group 2 

Group 3 

Group 4 

Group 5 

1960-1970 

45 54 70 80 

41 50 67 78 

38 45 63 75 

38 45 63 75 

38 45 63 75 

ASTM, A572 Group 1 

Group 2 

Group 3 

Group 4 

Group 5 

1970 – 1997 47 58 62 75 

48 58 64 75 

50 57 67 77 

49 57 70 81 

50 55 79 84 

A36 and Dual Grade 50 

Group 1 

Group 2 

Group 3 

Group 4 

1990 – 1997 

48 55 66 73 

48 58 67 75 

52 57 72 76 

50 54 71 76 

Notes: 1. Lower bound values are mean - two standard deviations, from statistical data. 

2. Expected values are mean values from statistical data. 

3. For wide-flange shapes, produced prior to 1997, indicated values are representative of material 

extracted from the web of the section. 

4. For structural plate, expected strength may be taken as 125% of the minimum specified value. 

Lower-bound strength should be taken as the minimum specified value. 

2.5.2 Component Properties 

Behavior of components, including beams and columns, is dictated by such properties as 

area, width-to-thickness and slenderness ratios, lateral torsional buckling resistance, and 

connection details. Component properties of interest are: 

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• original cross-sectional shape and physical dimensions, 

• size and thickness of additional connected materials, including cover plates, bracing, and 

stiffeners, 

• existing cross-sectional area, section moduli, moments of inertia, and torsional properties at 

critical sections, 

• as-built configuration of intermediate, splice, end, and base-plate connections, 

• current physical condition of base metal and connector materials, including presence of 

deformation. 

When performing detailed evaluations and loss estimates it is necessary to conduct a structural 

analysis of the building’s response to ground motion. Each of these properties is needed to 

characterize building performance in the seismic analysis. The starting point for establishing 

component properties should be the construction documents. Preliminary review of these 

documents should be performed to identify primary vertical- and lateral-load-carrying elements 

and systems, and their critical components and connections. In the absence of a complete set of 

building drawings, the design professional must obtain the necessary information on section and 

connection properties through a program of field investigation. 

2.5.3 Condition Assessment 

A condition assessment of the existing building and site conditions should be performed as 

part of the seismic evaluation process, regardless of the type of evaluation being performed. The 

goals of this assessment are: 

• To examine the physical condition of primary and secondary components and the presence of 

any degradation. 

• To verify or determine the presence and configuration of components and their connections, 

and the continuity of load paths between components, elements, and systems. 

• To review other conditions such as neighboring buildings and the presence of nonstructural 

components that may significantly influence building performance. 

The physical condition of existing components and elements, and their connections, must be 

examined for presence of degradation. Degradation may influence environmental effects (e.g., 

corrosion, fire damage, chemical attack) or past or current loading effects (e.g., overload, damage 

from past earthquakes, fatigue, fracture). The condition assessment should also examine for 

configuration problems observed in recent earthquakes, including effects of discontinuous 

components, improper welding, and poor fit-up. 

Component orientation, plumbness, and physical dimensions should be confirmed during an 

assessment. Connections in steel components, elements, and systems require special 

consideration and evaluation. The load path for the system must be determined, and each 

connection in the load path must be evaluated. This includes diaphragm-to-component and 

component-to-component connections. 

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The condition assessment also affords an opportunity to review other conditions that may 

influence steel elements and systems and overall building performance. Of particular importance 

is the identification of other elements and components that may contribute to or impair the 

performance of the steel system in question, including infills, neighboring buildings, and 

equipment attachments. Limitations posed by existing coverings, wall and ceiling space, infills, 

and other conditions should also be defined such that prudent rehabilitation measures may be 

planned. 

Commentary: In order to perform reliable performance assessments of buildings, 

it is important to have knowledge of the existing condition of the building and its 

components. However, the framing in most welded steel moment-frame (WSMF) 

buildings construction is protected from deterioration by fireproofing and other 

building finishes, and therefore, most WSMF buildings will remain in good 

condition throughout their service lives. Unless a WSMF building has been 

subjected to an extreme loading event, such as a fire, extreme windstorm, or 

strong earthquake, or the structure exhibits signs of deterioration, such as rust 

stains, or lack of plumb, exhaustive condition surveys of WSMF structures are not 

generally justified, except as required to confirm that the construction conforms 

to the available construction documents. 

2.5.3.1 Scope and Procedures 

The scope of a condition assessment should include all primary structural elements and 

components involved in gravity-load and lateral-load resistance. 

If coverings or other obstructions exist, indirect visual inspection through use of drilled holes 

and a fiberscope may be utilized. If this method is not appropriate, then local removal of 

covering materials may be necessary. The following guidelines should be used: 

• If detailed design drawings exist, exposure of at least one different primary connection should 

occur for each connection type. If no deviations from the drawings exist, the sample may be 

considered representative. If deviations are noted, then removal of additional coverings from 

primary connections of that type must be done until the design professional has adequate 

knowledge to continue with the evaluation and rehabilitation. 

• In the absence of construction drawings, the design professional should establish inspection 

protocols that will provide adequate knowledge of the building needed for reliable evaluation. 

Physical condition of components and connectors may also dictate the use of certain 

destructive and nondestructive test methods. If steel elements are covered by well-bonded 

fireproofing materials or encased in durable concrete, it is likely that their condition will be 

suitable. However, local removal effort is dictated by the component and element design. It may 

be necessary to expose more connections because of varying designs and the critical nature of the 

connections. 

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2.5.3.2 Quantifying Results 

The results of the condition assessment should be used in the preparation of building system 

analytical models for the evaluation of seismic performance. To aid in this effort, the results 

should be quantified and reduced with the following specific topics addressed: 

• component section properties and dimensions, 

• connection configuration and presence of any eccentricities, 

• type and location of column splices, and 

• interaction of nonstructural components and their involvement in lateral-load resistance. 

All deviations noted between available construction records and as-built conditions should be 

accounted for and considered in the structural analysis. 

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Steel Moment-Frame Buildings Chapter 3: Performance Evaluation 

3. Performance Evaluation 

3.1 Scope 

This chapter provides simplified criteria for evaluating the probable seismic performance of 

existing welded steel moment-frame buildings. These procedures may be used to quantify the 

ability of a building to achieve desired performance objectives, either before or after the 

construction of structural upgrades. It includes definition of performance objectives, discussions 

of expected performance of buildings conforming to FEMA-302 NEHRP Recommended 

Provisions for Seismic Regulations for New Buildings and Other Structures, and procedures for 

estimating a level of confidence that a building will provide a desired level of performance for 

specified earthquake hazards. It is applicable only to well configured, regular structures as 

defined in FEMA-302. A more detailed procedure, applicable to irregular structures and 

performance objectives based on deterministic earthquake scenarios is presented in Appendix A 

of these Recommended Criteria. 

Commentary: These recommendations only address methods of evaluating 

structural performance of welded steel moment-frame buildings. Although the 

performance of nonstructural components of buildings is critically important to 

the way in which buildings are used following an earthquake, treatment of this 

topic is beyond the scope of these Recommended Criteria. FEMA-273 – NEHRP 

Guidelines for the Seismic Rehabilitation of Buildings provides a more complete 

set of recommendations with regard to evaluating the performance of 

nonstructural components. 

FEMA-355F – State of the Art Report on Performance Prediction and 

Evaluation, presents, in detail, the basis for the procedures contained herein and 

the derivation of the various parameters used in the procedures. 

3.2 Performance Definitions 

The performance evaluation procedures contained in these Recommended Criteria permit 

estimation of a level of confidence that a structure will be able to achieve a desired performance 

objective. Each performance objective consists of the specification of a structural performance 

level and a corresponding hazard level, for which that performance level is to be achieved. For 

example, a seismic upgrade design may be intended to provide a 95% level of confidence that a 

structure provide Collapse Prevention or better performance for earthquake hazards with a 2% 

probability of exceedance in 50 years, or a 50% confidence level that a structure provide 

Immediate Occupancy or better performance, for earthquake hazards with a 50% probability of 

exceedance in 50 years. The user may determine the level of confidence associated with 

achieving any desired performance objective. 

Commentary: The performance evaluation procedures contained in these 

Recommended Criteria are based on an approach first developed in FEMA-273. 

However, substantial modifications have been made to the procedures presented 

in that document. 

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In FEMA-273, performance objectives are expressed in a deterministic 

manner. Each performance objective consists of the specification of a limiting 

damage state, termed a performance level, together with a specification of the 

ground motion intensity for which that (or better) performance is to be provided. 

This implies a warranty that if the specified ground motion is actually 

experienced by a building designed using the FEMA-273 procedures, damage will 

be no worse than that indicated in the performance objective. 

In reality, it is very difficult to predict with certainty how much damage a 

building will experience for a given level of ground motion. This is because there 

are many factors that affect the behavior and response of a building, such as the 

stiffness of nonstructural elements, the strength of individual building 

components, and the quality of construction, that can not be precisely defined and 

also, because the analysis procedures used to predict building response are not 

completely accurate. In addition, the exact character of the ground motion that 

will actually affect a building is itself very uncertain. Given these uncertainties, it 

is inappropriate to imply that performance can be predicted in an absolute sense, 

and correspondingly, that it is absolutely possible to produce designs that will 

achieve desired performance objectives. 

In recognition of this, these Recommended Criteria adopt a reliability-based 

probabilistic approach to performance evaluation that explicitly acknowledges 

these inherent uncertainties. These uncertainties are expressed in terms of a 

confidence level. If an evaluation indicates a high level of confidence, for 

example 90 or 95% that a performance objective can be achieved, then this means 

it is very likely (but not guaranteed) that the building will be capable of meeting 

the desired performance. If lower confidence is calculated, for example 50%, this 

is an indication that the building may not be capable of meeting the desired 

performance objective. If still lower confidence is calculated, for example 30%, 

then this indicates the building will likely not be able to meet the desired 

performance objective. Increased confidence in a building’s ability to provide 

specific performance can be obtained in three basic ways. 

• Providing the building with greater earthquake resistance, for example, by designing 

the structure to be stiffer and stronger 

• Reducing some of the uncertainty inherent in the performance evaluation process 

through the use of more accurate structural models and analyses and better data on 

the building’s configuration, strength and stiffness. 

• More accurately characterizing the uncertainties inherent in the performance 

evaluation process, for example, by using the more exact procedures of Appendix A of 

these Recommended Criteria. 

Refer also to the commentary in Section 3.2.1.2 for additional discussion of 

the probabilistic approach adopted by these Recommended Criteria. 

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3.2.1 Hazards 

3.2.1.1 General 

Earthquake hazards include direct ground fault rupture, ground shaking, liquefaction, lateral 

spreading, and land sliding. Of these various potential hazards, the one that effects the largest 

number of structures and causes the most widespread damage is ground shaking. Ground 

shaking is the only earthquake hazard that the structural design provisions of the building codes 

directly address. However, for structures located on sites where any of the other hazards can 

result in significant ground deformation, these hazards should also be considered in a structural 

performance evaluation. 

3.2.1.2 Ground Shaking 

Ground shaking hazards are typically characterized by a hazard curve, which indicates the 

probability that a given value of a ground motion parameter, for example peak ground 

acceleration, will be exceeded over a certain period of time, and by acceleration response spectra 

or ground motion accelerograms that are compatible with the values of the ground motion 

parameters obtained from the hazard curve and the local site geology. The ground shaking 

hazard maps provided with the FEMA-302 NEHRP Recommended Provisions and the FEMA273 NEHRP Rehabilitation Guidelines have been prepared based on hazard curves that have 

been developed by the United States Geological Survey for a grid-work of sites encompassing 

the United States and its territories. FEMA-302 defines two specific levels of hazard for 

consideration in design and specifies methods for developing response spectra for each of these 

levels. The two levels are: 

1. Maximum Considered Earthquake (MCE) ground shaking. This is the most severe level of 

ground shaking that is deemed appropriate for consideration in the design process for 

building structures, though not necessarily the most severe level of ground shaking that could 

ever be experienced at a site. In most regions, this ground shaking has a 2% probability of 

exceedance in 50 years, or roughly a 2,500 year mean recurrence interval. In regions of very 

high seismicity, near major active faults, the MCE ground shaking level is limited by a 

conservative, deterministic estimate of the ground shaking resulting from a maximum 

magnitude earthquake on the known active faults in the region. The probability that such 

deterministic ground shaking will be experienced at a site can vary considerably, depending 

on the activity rate of the individual fault. Refer to FEMA-303, Commentary to the NEHRP 

Recommended Provisions for Seismic Regulations for New Buildings and Other Structures 

for more detailed information on this issue. 

2. Design Earthquake (DE) ground shaking. This is the ground shaking level upon which 

design lateral forces, used as the basis for analysis and design in FEMA-302, are based. It is 

defined as a spectrum that is 2/3 of the shaking intensity calculated for the MCE spectrum, at 

each period. The probability that DE ground shaking will be experienced varies, depending 

on the regional, and, in some cases, site, seismicity. 

Commentary: The mean recurrence interval for Design Earthquake ground 

shaking will vary depending on regional and site seismicity. In areas of low 

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seismicity the hazard return period will generally range between 750-1,250 years 

and will remain relatively constant across neighboring communities. In areas of 

high seismicity the recurrence interval may range between 300-600 years and can 

vary significantly within a distance of a few miles. 

Performance evaluation conducted in accordance with these Recommended Criteria may be 

performed for any level of ground shaking. Ground shaking will typically be determined 

probabilistically, i.e., based on the probability that shaking of the specified intensity will be 

experienced at a site. Ground shaking must be characterized by an acceleration response 

spectrum or a suite of ground motion accelerograms compatible with that spectrum. In addition, 

a coefficient k that relates the rate of change in ground motion intensity with change in 

probability, is required. FEMA-273 provides guidelines for development of ground motion 

response spectra at different probabilities of exceedance. The procedures of this chapter use a 

default value for the coefficient k, as described in the commentary to Section 3.6. Performance 

evaluation for deterministic ground motion based on specific earthquake scenarios, for example 

an earthquake of given magnitude on a specific fault can also be performed. Appendix A 

provides procedures that may be used for deterministically defined hazards. 

Commentary: Detailed guidelines on ground-motion estimation and 

characterization are beyond the scope of this publication. Those interested in 

such information are referred to FEMA-303 and FEMA 274 Commentary to the 

NEHRP Guidelines for Seismic Rehabilitation of Buildings and references noted 

therein. 

Although Section 3.2 of these Recommended Criteria indicates that 

performance objectives are an expression of the desired performance for a 

building, given that ground motion of certain intensity is experienced, this is a 

significant simplification. In reality, the performance objectives are statements of 

the total probability that damage experienced by a building in a period of years 

will be more severe than the desired amount (performance level), given our 

knowledge of the site seismicity. Although it is transparent to the user, this is 

obtained by integrating the conditional probability that building response exceeds 

the limiting response for a performance level, given a ground motion intensity, 

over the probability of experiencing different intensities of ground motion, as 

represented by the site hazard curve, and specifically, the coefficient k which is 

the logarithmic slope of the hazard curve, at the desired hazard level. Thus, a 

performance objective that is stated as “meeting collapse prevention performance 

for ground shaking with a 2% probability of exceedance in 50 years” should 

more correctly be stated as being “less than a 2% chance in 50 years of damage 

more severe than the collapse prevention level, given the mean definition of 

seismicity.” 

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The procedures contained in this chapter neglect uncertainties associated 

with the definition of the seismicity, that is, the intensity of ground shaking at 

various probabilities. Such uncertainties can be as large, and perhaps larger, 

than the uncertainties associated with structural performance estimation. Thus 

the confidence calculated in accordance with the procedures of this chapter is 

really a confidence associate with structural performance, given the presumed 

seismicity. 

The simplified procedures presented in this chapter have been developed 

using hazard parameters typical of coastal California. They can be 

conservatively applied in regions of lower seismicity without the need to 

determine site specific hazard parameters. However, accurate definition of the 

hazard is a critical part of the performance evaluation procedures contained 

herein and in regions of lower seismicity, may result in calculation of higher 

confidence. Appendix A of these Recommended Criteria presents more detailed 

procedures that may be used to consider directly the site-specific characteristics 

of hazard in the evaluation of performance. 

3.2.1.3 Other Hazards 

In order to predict reliably the probable performance of a structure, it is necessary to 

determine if earthquake hazards other than ground shaking, including direct ground fault rupture, 

liquefaction, lateral spreading, and land sliding are likely to occur at a site and to estimate the 

severity of these effects. The severity of ground fault rupture, lateral spreading and land sliding 

is characterized by an estimate of permanent ground deformation. The severity of liquefaction is 

characterized by an estimate of the potential loss in bearing strength of subsoil layers and 

permanent ground settlement. In order to determine the performance of a structure which is 

subject to these hazards, the effects of the projected ground displacements should be evaluated 

using a mathematical model of the structure. The severity of these hazards (i.e. probability of 

exceedance) used in performance evaluation should be compatible with that used in specification 

of ground shaking hazards. 

Commentary: Most sites are not at significant risk from earthquake hazards 

other than ground shaking. However, these hazards can be very destructive to 

structures located on sites where they occur. Accurate determination of the 

propensity of a site to experience these hazards requires site-specific study by a 

competent earth scientist or geotechnical engineer. Guidelines on such 

assessments are beyond the scope of this publication. 

3.2.2 Performance Levels 

Building performance is a combination of the performance of both structural and 

nonstructural components. Table 3-1 describes the overall levels of structural and nonstructural 

damage that may be expected of buildings meeting two performance levels, termed Collapse 

Prevention and Immediate Occupancy. These performance descriptions are not precise and 

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variation among buildings must be expected within the same Performance Level. The structural 

performance levels are presented in Section 3.2.2.2. 

Table 3-1 Building Performance Levels 

Building Performance Levels 

Collapse Prevention Level Immediate Occupancy Level 

Overall Damage Severe Light 

General Little residual stiffness and strength, 

but gravity loads are supported. 

Large permanent drifts. Some exits 

may be blocked. Exterior cladding 

may be extensively damaged and 

some local failures may occur. 

Building is near collapse. 

Structure substantially retains 

original strength and stiffness. 

Minor cracking of facades, 

partitions, ceilings, and 

structural elements. Elevators 

can be restarted. Fire protection 

operable. 

Nonstructural components Extensive damage. Equipment and contents are 

generally secure, but may not 

operate due to mechanical 

failure or lack of utilities. 

Comparison with performance 

intended by FEMA-302 for 

SUG-I buildings when subjected 

to the Design Earthquake 

Significantly more damage and 

greater risk. 

Much less damage and lower 

risk. 

Comparison with performance 

intended by FEMA-302 for 

SUG-I buildings when subjected 

to the Maximum Considered 

Earthquake 

Same level of performance Much less damage and lower 

risk. 

SUG = Seismic Use Group 

Commentary: Building performance is expressed in terms of building 

performance levels. These building performance levels are discrete damage 

states selected from among the infinite spectrum of possible damage states that 

WMSF buildings could experience as a result of earthquake response. The 

particular damage states identified as building performance levels have been 

selected because these performance levels have readily identifiable consequences 

associated with the postearthquake disposition of the building that are meaningful 

to the building user community and also because they are quantifiable in 

technical terms. These include the ability to resume normal functions within the 

building, the advisability of postearthquake occupancy, and the risk to life safety. 

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Although a building’s performance is a function of the performance of both 

structural systems and nonstructural components and contents, only the structural 

performance levels are defined in these Recommended Criteria. The reference to 

nonstructural components above is to remind the reader of the probable 

performance of these elements at the various performance levels. 

3.2.2.1 Nonstructural Performance Levels 

These Recommended Criteria only addresses methods of evaluating structural performance of 

steel moment-frame buildings. Although the performance of nonstructural components of 

buildings are critically important to the way in which buildings are used following an earthquake, 

treatment of this topic is beyond the scope of these Recommended Criteria. FEMA-273 provides 

a more complete set of recommendations with regard to evaluating the performance of 

nonstructural components. 

3.2.2.2 Structural Performance Levels 

Two discrete structural performance levels, Collapse Prevention and Immediate Occupancy 

are defined in these Recommended Criteria. Table 3-2 relates these structural performance levels 

to the limiting damage states for framing elements of steel moment-frame structures. Acceptance 

criteria, which relate to the permissible interstory drifts and earthquake-induced forces for the 

various elements of steel moment-frame structures, are tied directly to these structural 

performance levels and are presented in later sections of these Recommended Criteria. 

Commentary: FEMA-273 defines three structural performance levels, Immediate 

Occupancy, Life Safety and Collapse Prevention and also defines two 

performance ranges. These performance ranges, rather than representing 

discrete damage states, span the entire spectrum of potential damage states 

between no damage and total damage. No acceptance criteria are provided for 

these performance ranges in FEMA-273. Rather, these must be determined on a 

project-specific basis, by interpolation or extrapolation from the criteria provided 

for the three performance levels. Performance ranges, as such, are not defined in 

these Recommended Criteria. However, compatible with the FEMA-273 

approach, users have the ability to create their own, custom performance levels, 

and to develop acceptance criteria for these levels, based on interpolation 

between the two performance levels, to suit the needs of a specific project. When 

such interpolation is performed, it is not possible to associate a confidence level 

with achievement of these intermediate performance definitions. 

The Life Safety performance level contained in FEMA-273 and FEMA-302 is 

not included in these Recommended Criteria. As defined in FEMA-273 and 

FEMA-302, the Life Safety level is a damage state in which significant damage 

has been sustained, although some margin remains against either partial or total 

collapse. In FEMA-273 this margin is taken as 1/3. That is, it is anticipated that 

a ground motion level that is 1/3 larger than that which results in the Life Safety 

performance level for a building would be required to bring the building to the 

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Collapse Prevention level. In FEMA-302, this margin is taken as 1⁄2, i.e. it is 

believed that buildings designed for Life Safety performance can experience 

approximately 50% greater motion before they reach the Collapse Prevention 

level. Due to the somewhat arbitrary definition of this performance level, and the 

fact that different guidelines and codes have selected alternative definitions for it 

(as described above), the Life Safety level has not been included in these 

Recommended Criteria. However, as with the performance ranges, users desiring 

to evaluate buildings for the Life Safety performance level may do so by 

interpolating between the acceptance criteria provided for the Collapse 

Prevention and Immediate Occupancy levels. 

Table 3-2 Structural Performance Levels 

Structural Performance Levels 

Elements Type Collapse Prevention Immediate Occupancy 

Girder Extensive distortion, local 

yielding and buckling. A few 

girders may experience partial 

fractures 

Minor local yielding and 

buckling at a few places. 

Column Moderate distortion; some 

columns experience yielding. 

Some local buckling of flanges 

No observable damage or 

distortion 

Beam-Column 

Connections 

Connection 

Type 11 

Some fractures with some 

connections experiencing near 

total loss of capacity 

Less than 10% of connections 

fractured on any one floor; 

minor yielding at other 

connections 

Connection 

Type 21 

Many fractures with some 

connections experiencing near 

total loss of capacity 

Less than 10% of connections 

fractured on any one floor; 

minor yielding at other 

connections 

Panel Zone Extensive distortion Minor distortion 

Column Splice No fractures No yielding 

Base Plate Extensive yielding of anchor 

bolts and base plate 

No observable damage or 

distortion 

Drift Interstory Large permanent Less than 1% permanent 

Notes: 1 Connection types are defined in Section 3.6.2.1, Table 3-9. 

3.2.2.2.1 Collapse Prevention Performance Level 

The Collapse Prevention structural performance level is defined as the postearthquake 

damage state in which the structure is on the verge of experiencing partial or total collapse. 

Substantial damage to the structure has occurred, potentially including significant degradation in 

the stiffness and strength of the lateral-force-resisting system, large permanent lateral 

deformation of the structure, and, to a more limited extent, degradation in the vertical loadcarrying capacity. However, all significant components of the gravity-load-resisting system must 

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continue to carry their gravity-load demands. The structure may not be technically or 

economically practical to repair and is not safe for re-occupancy; aftershock activity could 

credibly induce collapse. 

3.2.2.2.2 Immediate Occupancy Performance Level 

The Immediate Occupancy structural performance level is defined as the postearthquake 

damage state in which only limited structural damage has occurred. Damage is anticipated to be 

so slight that it would not be necessary to inspect the building for damage following the 

earthquake, and such little damage as may be present would not require repair. The basic 

vertical- and lateral-force-resisting systems of the building retain nearly all of their preearthquake strength and stiffness. The risk of life-threatening injury as a result of structural 

damage is very low. Buildings meeting this performance level should be safe for immediate 

postearthquake occupancy, presuming that damage to nonstructural components is suitably light 

and that needed utility services are available. 

Commentary: When a building is subjected to earthquake ground motion, a 

pattern of lateral deformations that varies with time is induced in the structure. 

At any given point in time, a particular state of lateral deformation will exist in 

the structure, and at some time within the period in which the structure is 

responding to the ground motion, a maximum pattern of deformation will occur. 

At relatively low levels of ground motion, the deformations induced within the 

building will be limited, and the resulting stresses that develop within the 

structural components will be within their elastic range of behavior. Within this 

elastic range, the structure will experience no damage. All structural components 

will retain their original strength, stiffness and appearance, and when the ground 

motion stops, the structure will return to its pre-earthquake condition. 

At more severe levels of ground motion, the lateral deformations induced in 

the structure will be larger. As these deformations increase, so will demands on 

the individual structural components. At different levels of deformation, 

corresponding to different levels of ground motion severity, individual 

components of the structure will be strained beyond their elastic range. As this 

occurs, the structure starts to experience damage in the form of buckling, yielding 

and fracturing of the various components. As components become damaged, they 

degrade in stiffness, and some elements will begin to lose their strength. In 

general, when a structure has responded to ground motion within this range of 

behavior, it will not return to its pre-earthquake condition when the ground 

motion stops. Some permanent deformation may remain within the structure and 

damage may be evident throughout. Depending on how far the structure has been 

deformed, and in what pattern, the structure may have lost a significant amount of 

its original stiffness and, possibly, strength. 

Brittle elements are not able to sustain inelastic deformations and will fail 

suddenly; the consequences may range from local and repairable damage to 

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collapse of the structural system. At higher levels of ground motion, the lateral 

deformations induced in a structure will (1), strain a number of elements to a 

point at which the elements degrade in stiffness and strength, or (2), as a result of 

P-D effects, the structure loses stability. Eventually, partial or total collapse of 

the structure can occur. 

The structural performance levels relate the extent of a building’s response to 

earthquake hazards to these various possible damage states. At the Immediate 

Occupancy Level, degradation of strength and stiffness in beam-column 

connections is limited to approximately 10% of the connections on any given floor 

and throughout the structure as a whole. The structure retains a significant 

portion of its original stiffness and most, if not all, of its strength, although some 

slight permanent drift may result. At the Collapse Prevention level, the building 

has experienced extreme damage. If laterally deformed beyond this point, the 

structure can experience instability and can collapse. 

3.3 Evaluation Approach 

The basic process of performance evaluation, as contained in these Recommended Criteria is 

to develop a mathematical model of the structure and to evaluate its response to the earthquake 

hazard by one or more methods of structural analysis. The structural analysis is used to predict 

the value of various structural response parameters. These include: 

• interstory drift, and 

• axial forces on individual columns. 

These structural response parameters are related to the amount of damage experienced by 

individual structural components as well as to the structure as a whole. For each performance 

level, these Recommended Criteria specify acceptance criteria (median estimates of capacity) for 

all the design parameters indicated above. Acceptability of structural performance is evaluated 

considering both local performance (at the element level) and global performance. Acceptance 

criteria have been developed on a reliability basis, incorporating demand and resistance factors 

related to the uncertainty inherent in the evaluation process and incorporating the variation 

inherent in structural response, such that a confidence level can be established with regard to the 

ability of a structure to provide specific performance at selected, low, probabilities of 

exceedance. 

Once an analysis is performed, predicted demands are adjusted by two factors, an analytical 

uncertainty factor ga, which corrects the analytically predicted demands for bias and uncertainty 

inherent in the analytical technique, and a demand variability factor, g, which accounts for other 

sources of variability in structural response. These predicted demands are compared against 

acceptance criteria, which have been modified by resistance factors f to account for uncertainties 

and variation inherent in structural capacity prediction. Procedures are given to calculate the 

level of confidence provided by a seismic evaluation or upgrade design, to achieve a specific 

performance objective, based on the ratio of factored demand to factored capacity. If the 

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predicted level of confidence is inadequate, then either more detailed investigations and analyses 

should be performed to improve the level of confidence attained with regard to performance, 

through the attainment of better understanding of the structure’s behavior and modification of the 

demand and resistance factors, or the structure should be upgraded such that a sufficient level of 

confidence can be attained given the level of understanding. If it is deemed appropriate to 

upgrade a structure to improve its probable performance, an iterative approach consisting of trial 

design, followed by verification analysis, evaluation of design parameters against acceptance 

criteria, and calculation of confidence level is repeated until an acceptable upgrade design 

solution is found. Procedures for estimating confidence are contained in Section 3.6. 

Commentary: These procedures adopt a demand and resistance factor design 

(DRFD) model for performance evaluation. This approach is similar to the Load 

and Resistance Factor design approach adopted by AISC LRFD except that the 

LRFD provisions are conducted on an element basis, rather than a structural 

system basis, and demands in these procedures can be drifts as well as forces and 

stresses. The purpose of this DRFD approach is to allow characterization of the 

confidence level inherent in a design with regard to a specific performance 

objective. 

The factored interstory drift demand, calculated from the analysis represents 

a median estimate of the probable maximum interstory drift demand, at the 

desired probability of exceedance. Tables in these Recommended Criteria 

provide interstory drift capacities for the two performance levels for regular, well 

configured structures, dependent on structural system and connection type, as 

well as resistance factors f, that adjust the estimated capacity of the structure to 

median values. Appendix A provides procedures for determination of f factors 

for connections for which project-specific qualification testing is performed and a 

procedure that may be used to determine interstory drift capacities for irregular 

structures. 

Once the factored demands and capacities are determined, an index 

parameter l is calculated from the ratio of the factored demands and capacities 

as indicated in Section 3.6. The value of l is then used to determine an 

associated confidence level based on tabulated values related to the uncertainty 

inherent in the estimation of the building’s demand and capacities. 

3.4 Analysis 

In order to evaluate the performance of a welded steel moment-frame building it is necessary 

to construct a mathematical model of the structure that represents its strength and deformation 

characteristics, and to conduct an analysis to predict the values of various design parameters 

when it is subjected to design ground motion. This section provides guidelines for selecting an 

appropriate analysis procedure and for modeling. General requirements for the mathematical 

model are presented in Section 3.5. 

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3.4.1 Alternative Procedures 

Four alternative analytical procedures are available for use in performance evaluation of 

welded steel moment-frame buildings. The basic analytical procedures are described in detail in 

FEMA-273. This section provides supplementary guidelines on the applicability of the FEMA273 procedures and also provides supplemental modeling recommendations. The four 

procedures are: 

• linear static procedure – an equivalent lateral force technique, similar, but not identical, to 

that contained in many model building code provisions, 

• linear dynamic procedure – an elastic, modal, response-spectrum analysis, 

• nonlinear static procedure – a simplified nonlinear analysis procedure in which the forces and 

deformations induced by a monotonically increasing lateral loading is evaluated using a 

series of incremental elastic analyses of structures that are sequentially degraded to represent 

the effects of structural nonlinearity, 

• nonlinear dynamic procedure – a nonlinear dynamic analysis procedure in which the response 

of a structure to a suite of ground motion histories is determined through numerical 

integration of the equations of motion for the structure. Structural stiffness is altered during 

the analysis to conform to nonlinear hysteretic models of the structural components. 

Commentary: The purpose of structural analyses performed as part of the 

performance evaluation process is to predict the values of key response 

parameters that are indicative of the structure’s performance when it is subjected 

to ground motion. Once the values of these response parameters are predicted, 

the structure is evaluated for adequacy (the appropriate level of confidence of 

achieving the desired performance) using the basic approach outlined in Section 

3.6. 

Analyses performed in support of design, as required by FEMA-302, evaluate 

the strength and deformation of the structure when it is subjected to a somewhat 

arbitrary level of loading. The loading is based on, but substantially reduced 

from, that predicted by an elastic analysis of the structure’s dynamic response to 

the expected ground motions. Specifically, the loading is reduced by a factor R to 

account approximately for the beneficial effects of inelastic response. 

Analyses conducted in support of performance evaluation, under these 

Recommended Criteria, take a markedly different approach. Rather than 

evaluating the forces and deformations induced in the structure under arbitrarily 

reduced loading levels, these analysis procedures attempt to predict, within 

probabilistically defined bounds, the actual values of the important response 

parameters in response to design ground motion. 

The ability of the performance evaluation to estimate reliably the probable 

performance of the structure is dependent on the ability of the analysis procedure 

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to predict the values of these response parameters within acceptable levels of 

confidence. The linear dynamic procedure is able to provide relatively reliable 

estimates of the response parameters for structures that exhibit elastic, or near 

elastic, behavior. The linear static procedure inherently has more uncertainty 

associated with its estimates of the response parameters because it accounts less 

accurately for the dynamic characteristics of the building. The nonlinear static 

procedure is more reliable than the linear procedures in predicting response 

parameters for buildings that exhibit significant nonlinear behavior, particularly 

if the buildings are irregular. However, it does not accurately account for the 

effects of higher mode response. If appropriate modeling is performed, the 

nonlinear dynamic approach is most capable of capturing the probable behavior 

of the building in response to ground motion. However, there are considerable 

uncertainties associated with the values of the response parameters predicted by 

this technique. 

3.4.2 Procedure Selection 

Table 3-3 indicates the recommended analysis procedures for various performance levels and 

conditions of structural regularity. 

3.4.3 Linear Static Procedure 

3.4.3.1 Basis of the Procedure 

Linear static procedure (LSP) analysis of steel moment-frame structures should be conducted 

in accordance with the recommendations of FEMA-273, except as noted herein. In this 

procedure, lateral forces are applied to the masses of the structure, and deflections and 

component forces under this applied loading are determined. Calculated internal forces typically 

will exceed those that the building can develop, because anticipated inelastic response of 

components and elements is not directly recognized by the procedure. The predicted interstory 

drifts and column axial forces are evaluated using the procedures of Section 3.6. 

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Table 3-3 Analysis Procedure Selection Criteria 

Structural Characteristics Analytical Procedure 

Performance 

Level 

Fundamental 

Period, T 

Regularity Ratio of Column 

to Beam Strength 

Linear 

Static 

Linear 

Dynamic 

Nonlinear 

Static 

Nonlinear 

Dynamic 

Immediate 

Occupancy 

T < 3.5Ts 

1 Regular or 

Irregular 

Any Conditions Permitted Permitted Permitted Permitted 

T > 3.5 Ts 

1 Regular or 

Irregular 

Any Conditions Not 

Permitted 

Permitted Not 

Permitted 

Permitted 

Collapse 

Prevention 

T < 3.5Ts 

1 Regular2 Strong Column3 Permitted Permitted Permitted Permitted 

Weak Column3 Not 

Permitted 

Not 

Permitted 

Permitted Permitted 

Irregular

2 Any Conditions Not 

Permitted 

Not 

Permitted 

Permitted Permitted 

T > 3.5Ts Regular Strong Column3 Not 

Permitted 

Permitted Not 

Permitted 

Permitted 

Weak Column3 Not 

Permitted 

Not 

Permitted 

Not 

Permitted 

Permitted 

Irregular2 Any Conditions Not 

Permitted 

Not 

Permitted 

Not 

Permitted 

Permitted 

Notes: 

1- Ts is the period at which the response spectrum transitions from a domain of constant response 

acceleration (the plateau of the response spectrum curve) to one of constant spectral velocity. Refer to 

FEMA-273 or FEMA-302 for more information 

2- Conditions of regularity are as defined in FEMA-273. These conditions are significantly different than 

those defined in FEMA-302. 

3- A structure qualifies as having a strong column condition if at every floor level, the quantity 

SMprc / SMprb is greater than 1.0, where SMprc and SMprb are the sum of the expected plastic moment 

strengths of the columns and beams that participate in the moment-resisting framing in a given direction of 

structural response. 

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Commentary: The linear static procedure is a method of estimating the response 

of the structure to earthquake ground shaking by representing the effects of this 

response through the application of a series of static lateral forces applied to an 

elastic mathematical model of the structure and its stiffness. The forces are 

applied to the structure in a pattern that represents the typical distribution of 

inertial forces in a regular structure responding in a linear manner to the ground 

shaking excitation, factored to account, in an approximate manner, for the 

probable inelastic behavior of the structure. It is assumed that the building 

response is dominated by the fundamental mode and that the lateral drifts 

induced, in the elastic structural model, by these forces represent a reasonable 

estimate of the actual deformation of the building when responding inelastically. 

In the LSP, the building is modeled with linearly-elastic stiffness and 

equivalent viscous damping that approximate values expected for loading to near 

the yield point. The static lateral forces, whose sum is equal to the pseudo lateral 

load, (so named in FEMA-273) represent earthquake demands for the LSP. The 

magnitude of the pseudo lateral load has been selected with the intention that 

when it is applied to the linearly elastic model of the building it will result in 

design displacement amplitudes approximating maximum displacements that are 

expected during the design earthquake. However, if the building responds 

essentially elastically to the design earthquake, the calculated internal forces will 

be reasonable approximations of those expected during the design earthquake. If 

the building responds inelastically to the design earthquake, as will commonly be 

the case, the internal forces that would develop in the yielding building will be 

less than the internal forces calculated on an elastic basis, but the predicted 

interstory drifts will approximate those that would actually occur in the structure. 

The performance of welded steel moment-frame buildings is most closely 

related to the total inelastic deformation demands on the various seismic elements 

that comprise the structure, such as plastic rotation demands on beam-column 

assemblies and tensile demands on column splices. Linear analysis methods do 

not permit direct evaluation of such demands. However, through a series of 

analytical evaluations of typical buildings for a number of earthquake records, it 

has been possible to develop statistical correlation between the interstory drift 

demands predicted by a linear analysis and the actual inelastic deformation 

demands determined by more accurate nonlinear methods. These correlation 

relationships are reasonably valid for regular buildings, using the definitions of 

regularity in FEMA-273. 

Although performance of welded steel moment-frame buildings is closely 

related to interstory drift demand, there are some failure mechanisms, notably, 

the failure of column splices, that are more closely related to strength demand. 

However, since inelastic structural behavior affects the strength demand on such 

elements, linear analysis is not capable of directly predicting these demands, 

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except when the structural response is essentially elastic. Therefore, when linear 

static analysis is performed for structures that respond in an inelastic manner, 

column axial demands should be estimated using a supplementary plastic analysis 

approach. 

The LSP is based on the assumption that the distribution of deformations 

predicted by an elastic analysis where all members remain linear elastic during 

all loadings, is similar to the distribution of deformations that will occur in actual 

nonlinear response. This assumption is inaccurate and can become more so for 

buildings that are highly irregular, that have response dominated by higher 

modes, or that experience large inelastic demands. It is for these reasons that 

alternative methods of analysis are recommended for irregular buildings and 

buildings with relatively long fundamental periods of vibration. 

3.4.3.2 Period Determination 

The fundamental period for each of the two orthogonal directions of building response shall 

be calculated by one of the following three methods. 

Method 1. Eigenvalue (dynamic) analysis of the mathematical model of the building. The 

model for buildings with flexible diaphragms shall consider representation of diaphragm 

flexibility unless it can be shown that the effects of omission will not be significant. 

Method 2. Evaluation of the following equation: 

0.8 T = Ct

hn (3-1) 

where

T = fundamental period (in seconds) in the direction under consideration,

Ct = 0.028 for steel moment frames,

hn = height (in feet) of the roof level above the base.

Method 3. The fundamental period of a one-story building with a single-span, flexible

diaphragm may be calculated as:

T = (0.1Dw + 0.078 ) Dd 

0.5 (3-2) 

where Dw and Dd are in-plane frame and diaphragm displacements, respectively, in inches, 

due to a lateral load, in the direction under consideration, equal to the weight tributary to the 

diaphragm. For multiple-span diaphragms, a lateral load equal to the gravity weight tributary 

to the diaphragm span under consideration should be applied to each diaphragm span to 

calculate a separate period for each diaphragm span. The loads from each diaphragm should 

then be distributed to the frames using tributary load assumptions. 

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Commentary: The approximate period formula indicated in Method 2 is different 

from that contained in either FEMA-273 or FEMA-302. This formula has been 

adapted from recent study of the statistical distribution of measured periods in 

buildings obtained from accelerometer recordings of excitation occurring in past 

earthquakes (Goel and Chopra, 1997). This formula is intended to provide 

approximately an 84% confidence level (mean+ 1 s) that the actual period will 

exceed the calculated value. The formula has intentionally been selected to 

underestimate the actual period of the building as this will result in a 

conservatively large estimate of the calculated pseudo lateral force applied to the 

structure as a loading (See Section 3.4.3.3.1). The large pseudo lateral force will 

result in conservatively large estimates of interstory drift. 

Use of the more accurate Method 1 procedure will typically result in lower 

estimates of interstory drift, and therefore increased confidence in the ability of a 

building to meet performance goals. 

3.4.3.3 Determination of Actions and Deformations 

3.4.3.3.1 Pseudo Lateral Load 

The pseudo lateral load, given by Equation 3-3, shall be independently calculated for each of 

the two orthogonal directions of building response, and applied to a mathematical model of the 

structure. 

V = C1C2 C3SaW (3-3) 

where: 

C1 = modification factor to relate expected maximum inelastic displacements to 

displacements calculated for linear elastic response. C1 may be calculated using 

the procedure indicated in Section 3.3.3.3 in FEMA-273 with the elastic base 

shear capacity substituted for Vy. Alternatively, C1 may be taken as having a value 

of 1.0 when the fundamental period T of the building response is greater than Ts 

and shall be taken as having a value of 1.5 when the fundamental period of the 

structure is equal to or less than T0. Linear interpolation shall be used to calculate 

C1 for intermediate values of T. 

T0 = period at which the acceleration response spectrum for the site reaches its peak 

value, as indicated in FEMA-302. It may be taken as 0.2Ts. 

TS = the characteristic period of the response spectrum, defined as the period associated 

with the transition from the constant spectral acceleration response segment of the 

spectrum to the constant spectral velocity response segment of the spectrum, as 

defined in FEMA-302. 

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C2 = a modification factor to represent the effect of hysteretic pinching on the 

maximum displacement response. For steel moment-frame structures the value of 

C2 shall be taken as 1.0. 

C3 = modification factor to represent increased dynamic displacements due to 

P-D effects and stiffness degradation. C3 may be taken from Table 3-4 or shall be 

calculated from the equation: 

C3 = 1+ 

5(qi - 0.1) ‡ 1.0 (3-4) T 

where: 

qi = the coefficient determined in accordance with Section 3.2.5.1 of 

FEMA-273. 

Sa = Response spectrum acceleration, at the fundamental period and damping ratio of 

the building in the direction under consideration, for the hazard level 

corresponding to the performance objective being evaluated (i.e., probability of 

exceedance). The value of Sa may be calculated using the procedure outlined in 

Section 2.6.1.5 of FEMA-273. 

W = Total dead load and anticipated live load as indicated below: 

• in storage and warehouse occupancies, a minimum of 25% of the floor live 

load, 

• the actual partition weight or minimum weight of 10 psf of floor area, 

whichever is greater, 

• the applicable snow load – see the NEHRP Recommended Provisions, 

• the total weight of permanent equipment and furnishings. 

Commentary: The pseudo lateral force, when distributed over the height of the 

linearly-elastic model of the structure, is intended to produce calculated lateral 

displacements approximately equal to those that are expected in the real structure 

during the design event. If it is expected that the actual structure will yield during 

the design event, the force given by Equation (3-3) may be significantly larger 

than the actual strength of the structure to resist this force. The acceptance 

evaluation procedures in Section 3.6 are developed to take this into account. 

The values of the C3 coefficient contained in Table 3-4 are conservative for 

most structures, and will generally result in calculation of an unduly low level of 

confidence. Use of Equation 3-4 to calculate C3 is one way to improve calculated 

confidence without extensive additional effort, and is recommended. 

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Table 3-4 Modification Factors C3 for Linear Static Procedure 

Performance Level C3 

Immediate Occupancy 1.0 

Collapse Prevention 

Type 11 FR connections 1.2 

Type 22 FR connections 1.4 

Notes: 

1. Type 1 connections are capable of resisting median total drift 

angle demands of 0.04 radians without fracture or strength 

degradation. 

2. Type 2 connections are capable of resisting median total drift 

angle demands of 0.01 radians without fracture or strength 

degradation. Generally, welded unreinforced connections, 

employing weld metal with low notch toughness, typical of older 

steel moment-frame buildings should be considered to be of this 

type. 

3.4.3.3.2 Vertical Distribution of Seismic Forces 

The lateral load Fx applied at any floor level x shall be determined as in Section 3.3.1.3B of 

FEMA-273. 

3.4.3.3.3 Horizontal Distribution of Seismic Forces 

The seismic forces at each floor level of the building shall be distributed according to the 

distribution of mass at that floor level. 

3.4.3.3.4 Diaphragms 

Floor and roof diaphragms shall be evaluated using the procedure outlined in Section 

3.3.1.3D in FEMA-273. The lateral seismic load on each flexible diaphragm shall be distributed 

along the span of that diaphragm, considering its displaced shape. 

3.4.3.3.5 Determination of Interstory Drift 

Interstory drifts shall be calculated using lateral loads calculated in accordance with Section 

3.4.3.3.1 and stiffness obtained from Section 3.5. Factored interstory drift demands gagdi at each 

story i, shall be determined by applying the appropriate analysis uncertainty factor ga and demand 

variability factor g obtained from Section 3.6.2. 

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Œ œ 

�

� � � 

�

� � � 

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Chapter 3: Performance Evaluation Steel Moment-Frame Buildings 

3.4.3.3.6 Determination of Column Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

calculated column forces by the applicable analysis uncertainty factor ga and demand variability 

factor g obtained in Section 3.6.3. Column forces shall be calculated either as: 

1. the axial demands from the unreduced linear analysis, or 

2. the axial demands computed from the equation: 

Ø � n M � � n M � ø 

P' c = –Œ2�

�� pe �

� - 2��� pe �� œ (3-5) 

o Ł i=x L łL Ł i=x L łR ß 

where: 

n M pe �

Ł i= x L �

�

ł L 

= the summation of the expected plastic moment strength (ZFye) divided by 

the span length L, of all moment-connected beams framing into the left 

hand side of the column, above the level under consideration, and 

n M pe �

Ł i= x L �

�

ł R 

= the summation of the expected plastic moment strength (ZFye) divided by 

the span length L, of all moment-connected beams framing into the right 

hand side of the column, above the level under consideration. 

When a column is part of framing that resists lateral forces under multiple directions of 

loading, the seismic demand shall be taken as the most severe condition resulting from 

application of 100% of the seismic demand computed for any one direction of response with 30% 

of the seismic demand computed for the orthogonal direction of response. 

3.4.4 Linear Dynamic Procedure 

3.4.4.1 Basis of the Procedure 

Linear dynamic procedure (LDP) analysis of steel moment frames shall be conducted in 

accordance with the recommendations in Section 3.3.2 of FEMA-273 except as specifically noted 

herein. Coefficients C1, C2, and C3 should be taken as indicated in Section 3.4.3.3 of these 

Recommended Criteria. 

Estimates of interstory drift and column axial demands shall be evaluated using the 

applicable procedures of Section 3.6. Calculated displacements and column axial demands are 

factored by the applicable analytical uncertainty factor ga and demand variability factor g obtained 

from Section 3.6 and compared with factored capacity values for the appropriate performance 

level. Calculated internal forces typically will exceed those that the building can sustain because 

of inelastic response of components and elements, but are generally not used to evaluate 

performance. 

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Commentary: The linear dynamic procedure is similar in approach to the linear 

static procedure, described in Section 3.4.3. However, because it directly 

accounts for the stiffness and mass distribution of the structure in calculating the 

dynamic response characteristics, it use introduces somewhat less uncertainty 

than does the LSP. Coefficients C1, C2, and C3, which account in an approximate 

manner for the differences between elastic predictions of response and inelastic 

behavior are the same as for the linear static method. Under the linear dynamic 

procedure, inertial seismic forces, their distribution over the height of the 

building, and the corresponding internal forces and system displacements are 

determined using a linearly elastic, response spectrum analysis. 

The basis, modeling approaches, and acceptance criteria of the LDP are 

similar to those for the LSP. The main exception is that the response calculations 

are carried out using modal response spectrum analysis (RSA). Modal spectral 

analysis is carried out using unreduced, linearly-elastic response spectra scaled 

to the hazard level (probability of exceedance) inherent in the desired 

performance objective. As with the LSP, it is expected that the LDP will produce 

estimates of displacements and interstory drifts that are approximately correct, 

but will produce estimates of internal forces that exceed those that would be 

obtained in a yielding building. 

3.4.4.2 Analysis 

3.4.4.2.1 General 

The LDP shall conform to the criteria in Section 3.3.2.2 of FEMA-273. The analysis shall be 

based on appropriate characterization of the ground motion. The requirement that all significant 

modes be included in the response analysis may be satisfied by including sufficient modes to 

capture at least 90% of the participating mass of the building in each of the building’s principal 

horizontal directions. Modal damping ratios should reflect the damping inherent in the building 

at deformation levels less than the yield deformation. Except for buildings incorporating passive 

or active energy dissipation devices, or base isolation technology, effective damping shall be 

taken as 5% of critical. 

The interstory drift, and other response parameters calculated for each mode, and required for 

evaluation in accordance with Section 3.4.4.3, should be combined by recognized methods to 

estimate total response. Modal combination by either the SRSS (square root of the sum of 

squares) rule or the CQC (complete quadratic combination) rule is acceptable. 

Multidirectional excitation effects may be accounted for by combining 100% of the response 

due to loading in direction A with 30% of the response due to loading in the direction B; and by 

combining 30% of the response in direction A with 100% of the response in direction B, where A 

and B are orthogonal directions of response for the building. 

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3.4.4.2.2 Ground Motion Characterization 

The horizontal ground motion should be characterized by one of the following methods: 

1. An elastic response spectrum, developed in accordance with the recommendations of Section 

2.6.1.5 in FEMA-273 for the hazard level contained in the desired performance objective. 

2. A site-specific response spectrum developed in accordance with the recommendations of 

Section 2.6.2.1 of FEMA-273 for the appropriate hazard level contained in the desired 

performance objective. 

3.4.4.3 Determination of Actions and Deformations 

3.4.4.3.1 Factored Interstory Drift Demand 

Factored interstory drift demand shall be obtained by multiplying the interstory drift results of 

the response spectrum analysis by the product of the modification factors, C1, C2, and C3 defined 

in Section 3.4.3 and by the applicable analytical uncertainty factor ga and demand variability 

factor g obtained from Section 3.6.2. 

3.4.4.3.2 Determination of Column Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

calculated column forces, as given in Section 3.4.3.3.6, by the applicable analysis uncertainty 

factor ga and demand variability factor g obtained from Section 3.6.3. 

3.4.5 Nonlinear Static Procedure 

3.4.5.1 Basis of the Procedure 

Under the nonlinear static procedure (NSP), a model directly incorporating the inelastic 

material and nonlinear geometric response is displaced to a target displacement, and resulting 

internal deformations and forces are determined. The nonlinear load-deformation characteristics 

of individual components and elements of the building are modeled directly. The mathematical 

model of the building is subjected to a pattern of monotonically increased lateral forces or 

displacements until either a target displacement is exceeded or mathematical instability occurs. 

The target displacement is intended to approximate the total maximum displacement likely to be 

experienced by the actual structure, at the hazard level corresponding to the selected performance 

objective. The target displacement should be calculated in accordance with the procedure 

presented in Section 3.3.3.3A of FEMA-273 with modifications, as indicated below. Because the 

mathematical model accounts directly for effects of material and geometric nonlinear response, 

the calculated internal forces will be reasonable approximations of those expected during the 

design earthquake, presuming that an appropriate pattern of loading has been applied. 

Interstory drifts and column axial demands obtained from the NSP are evaluated using the 

applicable procedures of Section 3.6. Calculated interstory drifts, column forces, and column 

splice forces are factored, and compared directly with factored acceptable values for the 

applicable performance level. 

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Commentary: The nonlinear static analysis approach inherently assumes that 

behavior is dominated by the first mode response of the structure. For this 

reason, this approach should be used only for structures with relatively short 

periods. What constitutes a building with a “short period” is dependent on the 

spectral characteristics of ground shaking anticipated at the site. The small 

magnitude events, that dominate the hazard at many central and eastern United 

States sites, tend to have most of their energy at very short periods, particularly 

on firm soil and rock sites. For sites subject to such shaking, nonlinear static 

analyses may be valid only for very short and rigid structures. The limitations on 

the use of NSP, based on period, contained in Table 3-3, are based on recent 

work that indicates that higher mode response does not tend to become significant 

in structures responding to ground shaking with typical response spectra unless 

the fundamental period of the structure is more than about 3.5 times the period at 

which the spectrum transitions from a range of constant acceleration response to 

constant velocity response. 

A second potential limitation of this procedure is that in practice, twodimensional models are often used to simulate three-dimensional response. 

Estimates of load distribution between the lateral-load-resisting elements in the 

building are required, and the accuracy of the analysis depends upon the 

accuracy of distribution. Three-dimensional linearly elastic models may be used 

to estimate this distribution; however, these models are unable to account for 

load redistribution occurring because of inelastic behavior. When many plastic 

hinges form nearly simultaneously, creating local frame mechanisms, the effects 

of torsional contributions may not be accurately represented. If a structure has 

significant torsional irregularity, three-dimensional models should be used. 

The NSP is also limited with regard to evaluation of simultaneous response to 

ground shaking in different directions. Little research has been performed on 

appropriate methods of accounting for multi-directional response using this 

technique. Therefore, these criteria have adapted standard approaches used in 

linear analysis for this purpose. 

3.4.5.2 Analysis Considerations 

3.4.5.2.1 General 

In the context of these Recommended Criteria, the NSP involves the application of 

incrementally adjusted, monotonically increasing lateral forces, or displacements, to a 

mathematical nonlinear model of a building, until the displacement of a control node in the 

mathematical model exceeds a target displacement. For buildings that are not symmetric about a 

plane perpendicular to the applied lateral loads, the lateral loads must be applied in both the 

positive and negative directions, and the maximum forces and deformations obtained from both 

directions used for design. 

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The relation between base shear force and lateral displacement of the control node should be 

established for control node displacements ranging between zero and 150% of the target 

displacement dt given by Equation 3-11 of FEMA-273. Performance evaluation shall be based on 

those column forces and interstory drifts corresponding to minimum horizontal displacement of 

the control node equal to the target displacement dt corresponding to the hazard level (probability 

of exceedance) appropriate to the performance objective being evaluated. 

Gravity loads shall be applied to appropriate elements and components of the mathematical 

model during the NSP. The loads and load combinations shall be as follows: 

1. 100% of computed dead loads and permanent live loads shall be applied to the model. 

2. 25% of transient floor live loads shall be applied to the model, except in warehouse and 

storage occupancies, where the percentage of live load used in the analysis shall be based on 

a realistic assessment of the average long-term loading. 

The analysis model should be discretized in sufficient detail to represent adequately the loaddeformation response of each component along its length. Particular attention should be paid to 

identifying locations of inelastic action along the length of a component, as well as at its ends. 

Commentary: As with any nonlinear model, the ability of the analyst to detect the 

presence of inelastic behavior requires the use of a nonlinear finite element at the 

assumed location of yielding. The model will fail to detect inelastic behavior when 

appropriately distributed finite elements are not used. However, as an alternative 

to the use of nonlinear elements, it is possible to use linear elements and 

reconfigure the model, for example, by adjusting member restraints, as 

nonlinearity is predicted to occur. For example, when a member is predicted to 

develop a plastic hinge, a linear model can be revised to place a hinge at this 

location. When this approach is used, the internal forces and stresses that caused 

the hinging must be reapplied as a nonvarying static load. 

The recommendation to continue the pushover analysis to displacements that 

are 150% of the target displacement is to allow an understanding of the probable 

behavior of the building under somewhat larger loading than anticipated. If the 

pushover analysis should become unstable prior to reaching 150% of the target 

displacement, this does not indicate that a design is unacceptable, but does 

provide an indication of how much reserve remains in the structure at the design 

ground motion. 

3.4.5.2.2 Control Node 

The NSP requires definition of a control node in the building. These Recommended Criteria 

consider the control node to be the center of mass at the roof of the building; the top of a 

penthouse should not be considered as the roof unless it is of such substantial area and 

construction as to materially affect the response. The displacement of the control node is 

compared with the target displacement – a displacement that characterizes the effects of 

earthquake shaking at the desired hazard level. 

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3.4.5.2.3 Lateral Load Patterns 

Lateral loads should be applied to the building in profiles given in Section 3.3.3.2C of 

FEMA-273. 

3.4.5.2.4 Period Determination 

The effective fundamental period Te in the direction under consideration shall be calculated 

using the force-displacement relationship of the NSP as described in Section 3.3.3.2D of FEMA273. 

3.4.5.2.5 Analysis of Three-Dimensional Models 

Static lateral forces shall be imposed on the three-dimensional mathematical model 

corresponding to the mass distribution at each floor level. 

Independent analysis along each principal axis of the three-dimensional mathematical model 

is permitted unless multidirectional evaluation is required by Section 3.2.7 in FEMA-273. Refer 

also to Section 3.4.5.3.4 of these Recommended Criteria. 

3.4.5.2.6 Analysis of Two-Dimensional Models 

Mathematical models describing the framing along each axis (axis 1 and axis 2) of the 

building should be developed for two-dimensional analysis. The effects of horizontal torsion 

should be considered as required by Section 3.2.2.2 of FEMA-273. 

3.4.5.2.7 Connection Modeling 

Existing, fully restrained, unimproved welded moment-resisting connections should be 

modeled as indicated in Section 6.2.1.2 of these Recommended Criteria. Simple shear tab 

connections with slabs present should be modeled as indicated in Section 6.2.2.1.2. Improved or 

upgraded fully restrained moment-resisting connections should be modeled as for unimproved 

connections except that the quantity qSD should be as indicated in Chapter 6 for the applicable 

connection type. 

3.4.5.3 Determination of Actions and Deformations 

3.4.5.3.1 Target Displacement 

The target displacement, dt, for buildings with rigid diaphragms at each floor level shall be 

estimated using the procedures of Section 3.3.3.3A of FEMA-273. Actions and deformations 

corresponding to the control node displacement equal to the target displacement shall be used for 

performance evaluation in accordance with Section 3.6. 

3.4.5.3.2 Diaphragms 

The lateral seismic load on each flexible diaphragm shall be distributed along the span of that 

diaphragm, considering its displaced shape. 

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3.4.5.3.3 Factored Interstory Drift Demand 

Factored interstory drift demand shall be obtained by multiplying the maximum interstory 

drift calculated at the target displacement by the applicable analytical uncertainty factor ga and 

demand variability factor g obtained from Section 3.6.2. 

3.4.5.3.4 Multidirectional Effects 

Multidirectional excitation effects may be accounted for by combining 100% of the response 

due to loading in direction A with 30% of the response due to loading in the direction B; and by 

combining 30% of the response in direction A with 100% of the response in direction B, where A 

and B are orthogonal directions of response for the building. 

An acceptable alternative to this approach is to perform the nonlinear static analysis 

simultaneously in two orthogonal directions by application of 100% of the loading in direction A 

simultaneously with 30% of the loading in direction B. Loading shall be applied until 100% of 

the target displacement in direction A is achieved. This procedure shall be repeated with 30% of 

the loading applied in direction A and 100% in direction B. 

3.4.5.3.5 Factored Column and Column Splice Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

calculated column forces at the target displacement by the applicable analytical uncertainty factor 

ga and demand variability factor, g, from Section 3.6.3. 

3.4.6 Nonlinear Dynamic Procedure 

3.4.6.1 Basis of the Procedure 

Under the Nonlinear Dynamic Procedure (NDP), inertial seismic forces, their distribution 

over the height of the building, and the corresponding internal forces and system displacements 

are determined using an inelastic response history dynamic analysis. 

The basis, the modeling approaches, and the acceptance criteria for the NDP are similar to 

those for the NSP. The main exception is that the response calculations are carried out using 

response-history analysis. With the NDP, the design displacements are not established using a 

target displacement, but instead are determined directly through dynamic analysis using suites of 

ground motion records. Calculated response can be highly sensitive to characteristics of 

individual ground motions; therefore, it is necessary to carry out the analysis with more than one 

ground motion record. Because the numerical model accounts directly for effects of material and 

geometric inelastic response, the calculated internal forces will be reasonable approximations of 

those expected during the design earthquake. 

Results of the NDP are to be checked using the applicable acceptance criteria of Section 3.6. 

Calculated displacements and internal forces are factored, and compared directly with factored 

acceptable values for the applicable performance level. 

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3.4.6.2 Analysis Assumptions 

3.4.6.2.1 General 

The NDP shall conform to the criteria given in Section 3.3.4.2A of FEMA-273. 

3.4.6.2.2 Ground Motion Characterization 

The earthquake shaking should be characterized by suites of ground motion acceleration 

histories, prepared in accordance with the recommendations of Section 2.6.2 of FEMA-273 and 

corresponding to the hazard level appropriate to the desired performance objective. A minimum 

of three pairs of ground motion records shall be used. Each pair shall consist of two orthogonal 

components of the ground motion. 

Consideration of multidirectional excitation effects required by Section 3.2.7 of FEMA-273 

may be satisfied by analysis of a three-dimensional mathematical model using simultaneously 

imposed pairs of earthquake ground motion records along the horizontal axes of the building. 

The effects of torsion should be considered according to Section 3.2.2.2 of FEMA-273. 

3.4.6.3 Determination of Actions and Deformations 

3.4.6.3.1 Response Quantities 

Response quantities shall be computed as follows: 

1. If less than seven pairs of ground motion records are used to perform the analyses, each 

response quantity (for example, interstory drift demand or column axial demand) shall be 

taken as the maximum value obtained from any of the analyses. 

2. If seven or more pairs of ground motion records are used to perform the analyses, the median 

value of each of the response quantities computed from the suite of analyses may be used as 

the demand. The median value shall be that value exceeded by 50% of the analyses in the 

suite. 

3.4.6.3.2 Factored Interstory Drift Demand 

Factored interstory drift demand shall be obtained by multiplying the maximum of the 

interstory drifts calculated in accordance with Section 3.4.6.3.1 by the applicable analytical 

uncertainty factor ga and demand variability factor g obtained from Section 3.6.2. 

3.4.6.3.3 Factored Column and Column Splice Demands 

Factored demands on columns and column splices shall be obtained by multiplying the 

column forces calculated in accordance with Section 3.4.6.3.1 by the applicable analytical 

uncertainty factor ga, and demand variability factor g from Section 3.6.3 or 3.6.4. 

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3.5 Mathematical Modeling 

3.5.1 Basic Assumptions 

In general, a steel moment-frame structure should be modeled and analyzed as a threedimensional assembly of elements and components. Although two-dimensional models may 

provide adequate response information for regular, symmetric structures and structures with 

flexible diaphragms, three-dimensional mathematical models should be used for analysis and 

design of buildings with plan irregularity as defined in FEMA-302. Two-dimensional modeling, 

analysis, and design of buildings with stiff or rigid diaphragms are acceptable, if torsional effects 

are either sufficiently small to be ignored, or are captured indirectly. 

Vertical lines of framing in buildings with flexible diaphragms may be individually modeled, 

analyzed and designed as two-dimensional assemblies of components and elements, or a threedimensional model may be used, with the diaphragms modeled as flexible elements. 

Explicit modeling of connection force-deformation behavior is not required for linear 

analysis procedures. In nonlinear procedures explicit modeling of connection stiffness is 

recommended for those cases when the connection is weaker than the connected components, or 

when it is appropriate to model strength degradation in the connection as a function of imposed 

deformation demand. 

Commentary: A finite element model will only collect information at places in the 

structure where a modeling element is inserted. When nonlinear deformations are 

expected in a structure, the analyst must anticipate the location of these 

deformations (such as plastic hinges) and insert nonlinear finite elements at these 

locations if the inelastic behavior is to be captured by the model. 

3.5.2 Frame Configuration 

The analytical model should accurately account for the stiffness of frame elements and 

connections. Element and component stiffness properties, strength estimates and locations of 

plastic hinge formation for both linear and nonlinear procedures can be determined from 

information given in Chapter 6 for typical connections. 

3.5.2.1 Elements Modeled 

Only the beams and columns forming the lateral-force-resisting system need be modeled, 

although it shall be permissible to model nonparticipating elements of the structure if realistic 

assumptions are made with regard to their stiffness, strength and deformation capacity. Refer to 

Chapter 6 for procedures for modeling common gravity-load beam-column connections. 

Commentary: Typically, engineers modeling steel moment-frame buildings 

neglect the participation of gravity-load-carrying beams and columns that are not 

intended to be part of the lateral-force-resisting system. Studies conducted in 

support of the development of these recommendations indicate that these 

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connections are capable of contributing non-negligible stiffness through large 

interstory drift demands. Analyses made with models that neglect the 

participation of these elements will tend to over-estimate demands on the lateralforce-resisting elements and interstory drift demand on the structure. 

While it is permissible to conduct performance evaluations using models that 

neglect non-participating framing, models that include the stiffness of these 

elements can be used to provide improved levels of confidence with regard to the 

building’s ability to meet desired performance objectives. This is an example of 

the process by which confidence can be improved, by performing more intense 

study to reduce the inherent uncertainty. 

3.5.2.2 Panel Zone Stiffness 

It shall be permissible for the model to assume centerline-to-centerline dimensions for the 

purpose of calculating stiffness of beams and columns. Alternatively, more realistic assumptions 

that account for the flexibility of panel zones may be used. Regardless, calculation of moments 

and shears should be performed at the face of the column. 

Commentary: Models that use centerline-to-centerline dimensions for calculation 

of beam and column stiffness tend to estimate conservatively the interstory drift 

demand on the structure. While it is permissible to conduct performance 

evaluations using models that neglect the effect of the panel zone on beam and 

column stiffness, models that include more realistic estimation of this effect can 

be used to provide improved levels of confidence with regard to the building’s 

ability to meet desired performance objectives. 

A number of models are available to represent panel zones of momentresisting connections. These range from simple models that idealize the panel 

zone as a scissors-type model that accounts explicitly for the shear stiffness of the 

panel zone, and to complex multi-element models that accounts both for shear 

stiffness of the panel zone and the effects of geometric distortion of the zone. 

Analyses of buildings using these various models reported in FEMA-355C 

indicate that the particular model used has relatively little impact on the 

predicted interstory drift demand. However, for nonlinear analysis models, the 

element selected to represent the panel zone can have significant impact on where 

plasticity in the structure is predicted to occur, e.g. in the panel zone itself, within 

the beam, or a combination of these regions. 

3.5.3 Horizontal Torsion 

The effects of actual horizontal torsion must be considered. In FEMA-302, the total torsional 

moment at a given floor level includes the following two torsional moments: 

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1. the actual torsion, that is, the moment resulting from the eccentricity between the centers 

of mass at all floors above and including the given floor, and the center of rigidity of the 

vertical seismic elements in the story below the given floor, and 

2. the accidental torsion, that is, an accidental torsional moment produced by horizontal 

offset in the centers of mass, at all floors above and including the given floor, equal to a 

minimum of 5% of the horizontal dimension at the given floor level measured 

perpendicular to the direction of the applied load. 

For the purposes of performance evaluation, under these Recommended Criteria, accidental 

torsion should not be considered. In buildings with diaphragms that are not flexible, the effect of 

actual torsion should be considered if the maximum lateral displacement dmax from this effect at 

any point on any floor diaphragm exceeds the average displacement davg by more than 10%. 

Commentary: Accidental torsion is an artificial device used by the building codes 

to account for actual torsion that can occur, but is not apparent in an evaluation 

of the center of rigidity and center of mass in an elastic stiffness evaluation. Such 

torsion can develop during nonlinear response of the structure if yielding 

develops in an unsymmetrical manner in the structure. For example if the frames 

on the east and west sides of a structure have similar elastic stiffness the structure 

may not have significant torsion during elastic response. However, if the frame 

on the east side of the structure yields significantly sooner than the framing on the 

west side, then inelastic torsion will develop. Rather than requiring that an 

accidental torsion be applied in the analysis, as do the building codes, these 

Recommended Criteria indirectly account for the uncertainty related to these 

torsional effects in the calculation of demand and resistance factors. 

3.5.4 Foundation Modeling 

In general, foundations may be modeled as unyielding. Assumptions with regard to the extent 

of fixity against rotation provided at the base of columns should realistically account for the 

relative rigidities of the frame and foundation system, including soil compliance effects, and the 

detailing of the column base connections. For purposes of determining building period and 

dynamic properties, soil-structure interaction may be modeled as permitted by the building code. 

Commentary: Most steel moment frames can be adequately modeled by assuming 

that the foundation provides rigid support for vertical loads. However, the 

flexibility of foundation systems (and the attachment of columns to those systems) 

can significantly alter the flexural stiffness at the base of the frame. Where 

relevant, these factors should be considered in developing the analytical model. 

3.5.5 Diaphragms 

Floor and roof diaphragms transfer earthquake-induced inertial forces to vertical elements of 

the seismic-force-resisting system. Connections between the edge beams of floor and roof 

diaphragms and vertical seismic framing elements must have sufficient strength to transfer the 

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maximum calculated diaphragm shear forces to the vertical framing elements. Requirements for 

evaluation of diaphragm components are given in Section 3.3 of FEMA-273. 

Development of the mathematical model should reflect the stiffness of the diaphragm. As a 

general rule, most floor slabs with concrete fill over metal deck may be considered to be rigid 

diaphragms and floors or roofs with plywood diaphragms should be considered flexible. The 

flexibility of unfilled metal deck, and concrete slab diaphragms with large openings should be 

considered in the analytical model. 

Mathematical models of buildings with diaphragms that are not rigid should be developed 

considering the effects of diaphragm flexibility. For buildings with flexible diaphragms at each 

floor level, the vertical lines of seismic framing may be designed independently, with seismic 

masses assigned on the basis of tributary area. 

3.5.6 P-D Effects 

P-D effects, caused by gravity loads acting on the displaced configuration of the building, 

may be critical in the seismic performance of steel moment-frame buildings, which are usually 

flexible and may be subjected to large lateral displacements. 

The structure should be investigated to ensure that lateral drifts induced by earthquake 

response do not result in a condition of instability under gravity loads. At each story, the quantity 

Yi should be calculated for each direction of response, as follows: 

Pdi i Y = (3-5 ) i Vyi hi 

where: 

Pi = portion of the total weight of the structure including dead, permanent live, and 

25% of transient live loads acting on all of the columns within story level i, 

Vyi = total plastic lateral shear force in the direction under consideration at story i, 

hi = height of story i, which may be taken as the distance between the centerline of 

floor framing at each of the levels above and below, the distance between the top 

of floor slabs at each of the levels above and below, or similar common points of 

reference, and 

di = lateral interstory drift in story i, from the analysis in the direction under 

consideration, at its center of rigidity, using the same units as for measuring hi. 

In any story in which Yi is less than or equal to 0.1, the structure need not be investigated 

further for stability concerns. When the quantity Yi in a story exceeds 0.1, the analysis of the 

structure should explicitly consider the geometric nonlinearity introduced by P-D effects. When 

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Yi in a story exceeds 0.3, the structure shall be considered unstable, unless a detailed global 

stability capacity evaluation for the structure, considering P-D effects, is conducted in accordance 

with the guidelines of Appendix A. 

For nonlinear procedures, second-order effects should be considered directly in the analysis; 

the geometric stiffness of all elements and components subjected to axial forces should be 

included in the mathematical model. 

Commentary: The values of interstory drift capacity for the Collapse Prevention 

performance level, provided in Section 3.6, and the corresponding resistance 

factors, were computed considering P-D effects (FEMA-355F). For a given 

structure, it is believed that if the value of Y is less than 0.3 the effects of P-D 

have been adequately considered by these general procedures. For values of Y 

greater than this limit the statistics on frame interstory drift capacities contained 

in Section 3.6 are inappropriate. For such frames explicit determination of 

interstory drift capacities, considering P-D effects using the detailed Performance 

Evaluation procedures outlined in Appendix A is required. 

The plastic story-shear quantity, Vyi should be determined by methods of 

plastic analysis. In a story in which: (1) all beam-column connections meet the 

strong-column-weak-beam criterion, (2) the same number of moment-resisting 

bays is present at the top and bottom of the frame, and (3) the strength of momentconnected girders at the top and bottom of the frame is similar, Vyi may be 

approximately calculated from the equation: 

n 

2� M pG j 

Vyi = 

j =1 (3-6) hi 

where: 

MpGj = the expected plastic moment capacity of each girder “j” 

participating in the moment resisting framing at the floor level on 

top of the story 

n = the number of moment-resisting girders in the framing at the floor 

level on top of the story 

In any story in which none of the columns meets the strong-column-weakbeam criterion, the plastic story-shear quantity, Vyi may be calculated from the 

equation: 

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n 

2� M pC k 

Vyi = 

k =1 

h (3-7) 

i 

where: 

MpCk = the plastic moment capacity of each column “k”, participating in 

the moment resisting framing, considering the axial load present 

on the column. 

For other conditions, the quantity Vyi must be calculated by plastic mechanism 

analysis, considering the vertical distribution of lateral forces on the structure. 

3.5.7 Multidirectional Excitation Effects 

Buildings should be evaluated for response due to seismic forces in any horizontal direction. 

For regular buildings, seismic displacements and forces may be assumed to act nonconcurrently 

in the direction of each principal axis of a building. For buildings with plan irregularity and 

buildings in which one or more components form part of two or more intersecting elements, 

multidirectional excitation effects should be considered, as indicated in Section 3.4 for the 

various analytical procedures. 

3.5.8 Vertical Ground Motion 

The effects of vertical excitation on horizontal cantilevers may be considered either by static 

or dynamic response methods. Vertical earthquake shaking may be characterized by a spectrum 

with ordinates equal to 2/3 of those of the horizontal spectrum unless alternative vertical 

response spectra are developed using site-specific analysis. Vertical earthquake effects on other 

beam elements and column elements need not be considered. 

Commentary: There is no evidence that response to vertical components of 

ground shaking has had any significant effect on the performance of steel 

moment-frame buildings. Consequently, the effect of this response is not 

recommended for consideration in the performance evaluation, except as required 

by the building code for the case of horizontal cantilever elements. 

Traditionally, vertical response spectra, when considered, have been taken as 

2/3 of the horizontal spectra developed for the site. While this is a reasonable 

approximation for most sites, vertical response spectra at near-field sites, located 

within a few kilometers of the zone of fault rupture, can have substantially 

stronger vertical response spectra than indicated by this approximation. 

Development of site-specific response spectra is recommended when vertical 

response must be considered for buildings on such sites. 

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3.6 Acceptance Criteria 

Acceptability of building performance should be evaluated by determining a level of 

confidence in the building’s ability to meet the desired performance objective(s). The parameters 

in Table 3-5 must be independently evaluated, using the procedures of Section 3.6.1 and the 

parameters and acceptance criteria of Sections 3.6.2, 3.6.3, and 3.6.4, for each performance 

objective evaluated. The controlling parameter is that which results in the calculation of the 

lowest confidence for building performance. 

Table 3-5 Performance Parameters Requiring Evaluation of Confidence 

Parameter Discussion 

Interstory drift The maximum interstory drift computed for any story of the structure shall be 

evaluated for global and local behavior (for Collapse Prevention and 

Immediate Occupancy). Refer to Section 3.6.2 

Column axial load The adequacy of each column to withstand its calculated maximum 

compressive demand shall be evaluated both for Collapse Prevention and 

Immediate Occupancy. Refer to Section 3.6.3 

Column splice tension The adequacy of column splices to withstand their calculated maximum 

tensile demands shall be evaluated both for Collapse Prevention and 

Immediate Occupancy. Refer to Section 3.6.4 

3.6.1 Factored-Demand-to-Capacity Ratio 

Confidence level is determined through evaluation of the factored-demand-to-capacity ratio 

given by the equation: 

l = 

gg aD (3-8) fC 

where: 

C = capacity of the structure, as indicated in Sections 3.6.2, 3.6.3, and 3.6.4, for interstory 

drift demand, column compressive demand and column splice tensile demand, 

respectively. 

D = calculated demand for the structure, obtained from structural analysis. 

g = a demand variability factor that accounts for the variability inherent in the prediction 

of demand related to assumptions made in structural modeling and prediction of the 

character of ground shaking as indicated in Sections 3.6.2, 3.6.3, and 3.6.4 for 

interstory drift demand, column compressive demand and column splice tensile 

demand, respectively. 

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ga = an analytical uncertainty factor that accounts for bias and uncertainty, inherent in the 

specific analytical procedure used to estimate demand as a function of ground shaking 

intensity, as indicated in Section 3.6.2, 3.6.3 and 3.6.4 for interstory drift demand, 

column compressive demand and column splice tensile demand, respectively. 

f = a resistance factor that accounts for the uncertainty and variability inherent in the 

prediction of structural capacity as a function of ground shaking intensity, as indicated 

in Section 3.6.2, 3.6.3 and 3.6.4 for interstory drift demand, column compressive 

demand and column splice tensile demand, respectively. 

l = a confidence index parameter from which a level of confidence can be obtained. See 

Table 3-6. 

Factored-demand-to-capacity ratio l shall be calculated using equation 3-8 for each of the 

performance parameters indicated in Table 3-5, which also references the appropriate section of 

these Recommended Criteria where the various parameters, ga, g, f required to perform this 

evaluation may be found. These referenced sections also define an uncertainty parameter bUT 

associated with the evaluation of global and local interstory drift capacity, column compressive 

capacity, and column splice tensile capacity, respectively. These uncertainties are related to the 

building’s configuration, the type of moment-resisting connections present (type 1 or type 2), the 

type of analytical procedure employed and the performance level being evaluated. Table 3-6 

indicates the level of confidence associated with various values of the factored-demand-tocapacity ratio l calculated using Equation 3-8, for various values of the uncertainty parameter 

bUT. Linear interpolation between the values given in Table 3-6 may be used for values of 

factored-demand-to-capacity ratio l and uncertainty bUT intermediate to those tabulated. 

Table 3-6 Factored-Demand-to-Capacity Ratios l for Specific Confidence Levels and 

Uncertainty bUT factors 

Confidence Level 

Uncertainty 

Parameter bUT 

10 20 30 40 50 60 70 80 90 95 99 

0.2 1.37 1.26 1.18 1.12 1.06 1.01 0.96 0.90 0.82 0.76 0.67 

0.3 1.68 1.48 1.34 1.23 1.14 1.06 0.98 0.89 0.78 0.70 0.57 

0.4 2.12 1.79 1.57 1.40 1.27 1.15 1.03 0.90 0.76 0.66 0.51 

0.5 2.76 2.23 1.90 1.65 1.45 1.28 1.12 0.95 0.77 0.64 0.46 

0.6 3.70 2.86 2.36 1.99 1.72 1.48 1.25 1.03 0.80 0.64 0.43 

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Table 3-7 provides minimum recommended levels of confidence for each of the potential 

controlling behavior modes, that is, global stability, local connection capacity, column buckling 

or column splice tensile failure, for the Immediate Occupancy and Collapse Prevention 

performance levels, respectively. 

Table 3-7 Recommended Minimum Confidence Levels 

Behavior 

Performance Level 

Immediate Occupancy Collapse Prevention 

Global Behavior Limited by Interstory 

Drift 

50% 90% 

Local Connection Behavior Limited by 

Interstory Drift 

50% 50% 

Column Compression Behavior 50% 90% 

Column Splice Tension Behavior 50% 50% 

Commentary: In order to predict structural performance, these procedures rely 

on the application of structural analysis and laboratory test data to predict the 

behavior of real structures. However, there are a number of sources of 

uncertainty inherent in the application of analysis and test data to performance 

prediction. For example, the actual strength of structural materials, the quality of 

individual welded joints, and the amount of viscous damping present is never 

precisely known, but can have impact on both the actual amount of demand 

produced on the structure and its elements and their capacity to resist these 

demands. If the actual values of these and other parameters that affect structural 

performance were known, it would be possible to predict accurately both demand 

and capacity. However, this is never the case. In these procedures, confidence is 

used as a measure of the extent that predicted behavior is likely to represent 

reality. 

The extent of confidence inherent in a performance prediction is related to the 

possible variation in the several factors that affect structural demand and 

capacity, such as stiffness, damping, connection quality, and the analytical 

procedures employed. In this project, evaluations were made of the potential 

distribution of each of these factors and the effect of variation in these factors on 

structural demand and capacity. Each of these sources of uncertainty in 

structural demand and capacity prediction were characterized as part of the 

supporting research for this project, by a coefficient of variation, bU. The 

coefficient, bUT is the total coefficient of variation, considering all sources of 

uncertainty. It is used, together with other factors to calculate the demand and 

resistance factors. We assume that demand and capacity are lognormally 

distributed relative to these uncertainty parameters. This allows confidence to be 

calculated as a function of the number of standard deviations that factored3-36
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demand-to-capacity-ratio l lies above or below a mean value. Table 3-6 provides 

a solution for this calculation, using a conservative estimate of the hazard 

parameter, k=3.0, that is representative of the typical seismicity of coastal 

California. Further information on this method may be found in Appendix A. 

3.6.2 Performance Limited by Interstory Drift Angle 

3.6.2.1 Factored Interstory Drift Angle Demand 

Factored interstory drift demand should be computed as the quantity ggaD where the demand 

D, is the largest interstory drift in any story computed from structural analysis, ga is the 

coefficient obtained from Table 3-8, and g is the coefficient obtained from Table 3-9. 

Table 3-8 Interstory Drift Angle Analysis Uncertainty Factors, ga 

Analysis Procedure LSP LDP NSP NDP 

System Characteristic I.O.1 C.P.2 I.O.1 C.P.2 I.O.1 C.P.2 I.O.1 C.P.2 

Type 1 Connections 

Low Rise (<4 stories) 0.94 0.70 1.03 0.83 1.13 0.89 1.02 1.03 

Mid Rise (4-12 stories) 1.15 0.97 1.14 1.25 1.45 0.99 1.02 1.06 

High Rise (> 12 stories) 1.12 1.21 1.21 1.14 1.36 0.95 1.04 1.10 

Type 2 Connections 

Low Rise (<4 stories) 0.79 0.98 1.04 1.32 0.95 1.31 1.02 1.03 

Mid Rise (4-12 stories) 0.85 1.14 1.10 1.53 1.11 1.42 1.02 1.06 

High Rise (> 12 stories) 0.80 0.85 1.39 1.38 1.36 1.53 1.04 1.10 

Notes: 1- I.O. = Immediate Occupancy Performance Level 

2- C.P. = Collapse Prevention Performance Level 

Commentary: Several structural response parameters are used to evaluate 

structural performance. The primary parameter is interstory drift. Interstory 

drift is an excellent parameter for judging the ability of a structure to resist P-D 

instability and collapse. It is also closely related to plastic rotation demand, or 

drift angle demand, on individual beam-column connection assemblies, and is 

therefore a good predictor of the performance of beams, columns and 

connections. For tall slender structures, a significant portion of interstory drift is 

a result of axial elongation and shortening of different rows of columns. 

Although modeling of the structure should account for this frame flexibility, that 

portion of interstory drift resulting from axial column deformation in stories 

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below the story under consideration should be neglected in determining local 

connection performance. This portion of the interstory drift must usually be 

determined manually as most computer programs do not calculate this quantity 

separately. 

Table 3-9 Interstory Drift Angle Demand Variability Factors g 

Building Height 

Immediate 

Occupancy 

(I.O.) 

Collapse Prevention 

(C.P.) 

Type 1 Connections1 

Low Rise 

(3 stories or less) 

1.5 1.3 

Mid Rise (4-12 stories) 1.4 1.2 

High Rise 

(more than 12 stories) 

1.4 1.5 

Type 2 Connections2 

Low Rise 

(3 stories or less) 

1.4 1.4 

Mid Rise (4-12 stories) 1.3 1.5 

High Rise 

(more than 12 stories) 

1.6 1.8 

Notes: 

1- Type 1 connections are capable of resisting median total drift angle demands of 0.04 

radians without fracture or strength degradation. 

2- Type 2 connections are capable of resisting median total drift angle demands of 0.01 

radians without fracture or strength degradation. Generally, welded unreinforced 

connections, employing weld metal with low notch toughness, typical of older welded 

steel moment-frame buildings should be considered to be this type. 

3.6.2.2 Factored Interstory Drift Angle Capacity 

Interstory drift capacity may be limited either by the global response of the structure, or by 

the local behavior of beam-column connections. Section 3.6.2.2.1 provides values for global 

interstory drift capacity for regular, well-configured structures as well as associated uncertainties, 

bUT. Global interstory drift capacities for irregular structures must be determined using the 

detailed procedures of Appendix A. Section 3.6.2.2.2 provides procedures for evaluating local 

interstory drift angle capacity, as limited by connection behavior. 

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3.6.2.2.1 Global Interstory Drift Angle 

Factored interstory drift angle capacity, fC, as limited by global response of the building, 

shall be based on the product of the resistance factor f and capacity C, which are obtained from 

Table 3-10, for either Type 1 or Type 2 connections. Type 1 connections are capable of resisting 

median total interstory drift angle demands of 0.04 radians without fracturing or strength 

degradation. Type 2 connections are capable of resisting median total interstory drift angle 

demands of 0.01 radian without fracturing or strength degradation. Welded unreinforced 

moment-resisting connections with weld metal with low notch toughness should be considered 

Type 2. Table 3-11 provides values of the uncertainty coefficient bUT to be used with global 

interstory drift evaluation. 

Table 3-10 Global Interstory Drift Angle Capacity C and Resistance Factors f for 

Regular Buildings 

Building Height Performance Level 

Immediate Occupancy Collapse Prevention 

Interstory 

Drift Angle 

Capacity C 

Resistance 

Factor f 

Interstory 

Drift Angle 

Capacity C 

Resistance 

Factor f 

Type 1 Connections 

Low Rise (3 stories or less) 0.02 1.0 0.10 0.90 

Mid Rise ( 4 – 12 stories) 0.02 1.0 0.10 0.85 

High Rise (> 12 stories) 0.02 1.0 0.085 0.75 

Type 2 Connections 

Low Rise (3 stories or less) 0.01 1.0 0.10 0.85 

Mid Rise ( 4 – 12 stories) 0.01 0.9 0.08 0.70 

High Rise (> 12 stories) 0.01 0.85 0.06 0.60 

3.6.2.2.2 Local Interstory Drift Angle 

Factored interstory drift angle fC limited by local connection response, shall be based on the 

capacity C of the connection and resistance factor f obtained from Chapter 6 of these 

Recommended Criteria. For Immediate Occupancy performance, capacity C shall be taken as the 

quantity qIO and for Collapse Prevention performance, the quantity qU indicated in Chapter 6 for 

the connection types present in the building. For connection types not include in Chapter 6, the 

capacity and resistance factors should be obtained from laboratory testing and the procedures of 

Appendix A. Table 3-12 provides values of the uncertainty coefficient bUT for use in evaluating 

performance as limited by local connection behavior. 

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Table 3-11 Uncertainty Coefficient bUT for Global Interstory Drift Evaluation 

Building Height Performance Level 

Immediate Occupancy Collapse Prevention 

Type 1 Connections 

Low Rise (3 stories or less) 0.20 0.3 

Mid Rise ( 4 – 12 stories) 0.20 0.4 

High Rise (> 12 stories) 0.20 0.5 

Type 2 Connections 

Low Rise (3 stories or less) 0.20 0.35 

Mid Rise ( 4 – 12 stories) 0.20 0.45 

High Rise (> 12 stories) 0.20 0.55 

Notes: 1- Value of bUT should be increased by 0.05 for LSP analysis 

2- Value of bUT may be reduced by 0.05 for NDP analysis 

Table 3-12 Uncertainty Coefficient bUT for Local Interstory Drift Evaluation 

Building Height Performance Level 

Immediate Occupancy Collapse Prevention 

Type 1 Connections 

Low Rise (3 stories or less) 0.30 0.30 

Mid Rise ( 4 – 12 stories) 0.30 0.35 

High Rise (> 12 stories) 0.30 0.40 

Type 2 Connections 

Low Rise (3 stories or less) 0.30 0.35 

Mid Rise ( 4 – 12 stories) 0.30 0.40 

High Rise (> 12 stories) 0.30 0.40 

Notes: 1- Value of bUT should be increased by 0.05 for LSP analysis 

2- Value of bUT may be reduced by 0.05 for NDP analysis 

3.6.3 Performance Limited by Column Compressive Capacity 

3.6.3.1 Column Compressive Demand 

Factored column compressive demand shall be determined for each column as the quantity 

ggaD, where: 

D = the compressive axial load on the column determined as the sum of Dead Load, 25% of 

unreduced Live Load, and Seismic Demand. Seismic demand shall be determined by 

one of the following four analysis methods: 

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g

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Linear: The axial demands may be taken as those predicted by a linear 

static or linear dynamic analysis, conducted in accordance with 

Section 3.4.3 or 3.4.4. 

Plastic: The axial seismic demands may be taken from plastic analysis, as 

indicated by Equation 3-5 in Section 3.4.3.3.6. 

Nonlinear Static: The axial demands may be taken from the computed forces from a 

nonlinear static analysis, at the target displacement, in accordance 

with Section 3.4.5. 

Nonlinear Dynamic: The axial demands may be taken from the computed design forces 

from a nonlinear dynamic analysis, in accordance with Section 

3.4.6. 

ga = analytical uncertainty factor, taken from Table 3-13. 

g = demand variability demand factor, taken as having a value of 1.05. 

The uncertainty coefficient bUT shall be taken as indicated in Table 3-13 based on the 

procedure used to calculate column compressive demand D. 

Table 3-13 Analysis Uncertainty Factor ga and Total Uncertainty Coefficient bUT for 

Evaluation of Column Compressive Demands 

Analytical Procedure Analysis Uncertainty 

Factor 

ga 

Total Uncertainty 

Coefficient bUT 

Linear Static or Dynamic Analysis 1.15 0.35 

Plastic Analysis (Section 3.4.3.3.6) 1.0 0.15 

Nonlinear Static Analysis 1.05 0.20 

Nonlinear Dynamic Analysis 2 1 .4b

e 2 0 .0225 + b

Note: b may be taken as the coefficient of variation (COV) of the axial load values determined from the suite of 

nonlinear analyses 

Commentary: The value of g has been computed assuming a coefficient of 

variation for axial load values resulting from material strength variation and 

uncertainty in dead and live loads of 15%. The values of ga have been calculated 

assuming coefficients of variation of 30%, 0% and 15%, related to uncertainty in 

the analysis procedures for linear, plastic and nonlinear static analyses, 

respectively. In reality, for structures that are stressed into the inelastic range, 

elastic analysis will typically overestimate axial column demands, in which case, 

a value of 1.0 could be used. However, for structures that are not loaded into the 

inelastic range, the indicated value is appropriate. Plastic analysis will also 

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typically result in an upper bound estimate of column demand and application of 

additional demand factors is not appropriate. For nonlinear dynamic analysis, 

using a suite of ground motions, direct calculation of the analysis demand factor 

is possible, using the equation shown. All of these demand factors are based on 

the hazard parameter k having a value of 3, which is typical of conditions in 

coastal California. 

3.6.3.2 Column Compressive Capacity 

Factored compressive capacity of each individual column to resist compressive axial loads 

shall be determined as the product of the resistance factor, f, and the nominal axial strength C of 

the column, which shall be determined in accordance with the AISC Load and Resistance Factor 

Design Specification. Specifically, for the purposes of this evaluation, the effective length 

coefficient k shall be taken as having a value of 1.0 and the resistance factor f shall be assigned a 

value of 0.95. 

3.6.4 Column Splice Capacity 

The capacity of column tensile splices, other than splices consisting of complete joint 

penetration (CJP) butt welds of all elements of the column (flanges and webs) shall be evaluated 

in accordance with this section. Column splices consisting of CJP welds of all elements of the 

column, and in which the weld filler metal conforms to the recommendations of Sections 6.4.2.4 

and 6.4.2.5 of these Recommended Criteria need not be evaluated. 

3.6.4.1 Column Splice Tensile Demand 

Factored column splice tensile demand shall be determined for each column as the quantity 

ggaD where D is the column splice tensile demand. Column splice tensile demand shall be 

determined as the computed Seismic Demand in the column less 90% of the computed Dead 

Load demand. Seismic Demand shall be as determined for column compressive demand, in 

accordance with Section 3.6.3.1. The demand variability factor g shall be taken as having a value 

of 1.05 and the analysis uncertainty factor ga shall be taken as indicated in Table 3-13. The total 

uncertainty coefficient bUT shall also be taken as indicated in Table 3-13. 

3.6.4.2 Column Splice Tensile Capacity 

The capacity of individual column splices to resist tensile axial loads shall be determined as 

the product of the resistance factor, f, and the nominal tensile strength of the splice, C, as 

determined in accordance with the AISC Load and Resistance Factor Design Specification. 

Specifically, Chapter J of the AISC Specification shall be used to calculate the nominal tensile 

strength of the splice connection. For the purposes of this evaluation, f shall be assigned a value 

of 0.9. 

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Steel Moment-Frame Buildings Chapter 4: Loss Estimation 

4. Loss Estimation 

4.1 Scope 

This chapter provides data that may be used to perform estimates of probable repair costs for 

steel moment-frame buildings based on actuarial data obtained from the 1994 Northridge 

earthquake. These data may be used to estimate the cost of repair for these buildings, within 

levels of confidence, given limited data on the building characteristics and an estimate of ground 

motion intensity. A more detailed approach that incorporates the information obtained from a 

structural analysis of the building is contained in Appendix B of these Recommended Criteria. 

When an earthquake damages a building, there are a number of potential sources of economic 

loss. One source of losses is the cost associated with repairing the damage and restoring the 

building to service. Such losses are known as direct loss. Other sources of economic loss can 

result from an inability to occupy space in the damaged structure until it is repaired, the need to 

rent space for temporary or alternative quarters, relocation costs, litigation, devaluation of 

property values and a general decline in the economic welfare of the affected region. These 

losses are generally termed indirect losses. These Recommended Criteria provide methods only 

for estimating direct losses due to earthquake ground shaking. 

The direct losses that can be estimated using the methods of these Recommended Criteria 

typically represent only a small portion of the total losses caused by earthquakes. The other 

indirect losses are a function of a number of complex factors that relate to the economic and 

social makeup of the affected region, and the decision making process performed by individual 

owners and tenants and go far beyond the considerations of damage sustained by individual 

buildings. Therefore, although such losses are very important, they are considered to be beyond 

the scope of these Recommended Criteria. 

Although the tools presented herein can be applied to building specific loss estimates, they 

were originally developed with the intent of application to broad populations of buildings. The 

estimation of losses that may occur to a specific building in future earthquakes of unknown 

source, magnitude and distance is fraught with great uncertainty. Users are cautioned that actual 

performance of specific buildings in response to specific earthquake demands can be 

substantially different from what would be suggested by the statistically based methods presented 

in this chapter. 

4.2 Loss Estimation Methods 

Two alternative methods are provided in these Recommended Criteria to estimate probable 

repair costs for buildings due to future earthquake ground motion. The Rapid Loss Estimation 

Method, contained in this chapter, provides estimates of losses as a function of basic information 

about the building and estimates of seismic demands. The Detailed Loss Estimation Method 

found in Appendix B, directly utilizes engineering data obtained from a detailed structural 

analysis of the specific building. This Detailed Loss Estimation Method is compatible with 

FEMA’s HAZUS (NIBS, 1997a,b) loss estimation software and can be used to generate building4-1
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specific vulnerability information for use with that system and other similar loss estimation 

models. 

Commentary: The most common bases for producing loss estimates may be 

classified as historical experience, expert opinion, and engineering. All of these 

methods include significant uncertainty with regard to predicted damage and 

repair costs. 

Historical experience-based estimates are developed based on statistics on the 

actual damage and costs incurred for given classes of structures subjected to 

estimated or recorded seismic demands. When such data are available, it is 

possible to determine the distribution of losses over the population contained in 

the database, including a median (best estimate of the loss for any structure in the 

class) and a measure of variation. This permits the loss for a structure similar to 

those in the database to be estimated within a range of confidence. Significant 

sources of uncertainty include the lack of database completeness, differences 

between the structure being evaluated and the general population in the database, 

and the seismic demand range captured in the database. No database is 

comprehensive. 

The most commonly used loss estimation methodology is based on expert 

opinion of probable repair costs. ATC-13 (ATC, 1985) and other similar studies 

have developed damage functions by obtaining opinions from structural engineers 

and other experts on typical levels of damage for various classes of structures 

when subjected to different intensities of ground motion. Statistical data from 

such opinion surveys can then be used to derive loss estimates for other buildings. 

This approach also has much uncertainty and little to no direct tie to actual losses 

experienced in past events, other than as perceived by the experts. The 1994 

Northridge earthquake illustrates the uncertainty inherent in expert opinion, in 

that the brittle fractures that occurred in steel moment-frame buildings had not 

previously been anticipated. 

The Rapid Loss Estimation Method presented here uses both historical 

experience and expert opinion. A database of steel moment-frame building 

damage caused by the 1994 Northridge earthquake represented the historical 

experience. This was augmented by expert opinion where actuarial data were 

sparse or unreliable. 

The Detailed Loss Estimation method uses engineering calculations to 

estimate the types of damage likely to be experienced by the structure. Probable 

repair costs are then determined based on this damage. Such an approach has 

not been widely used in the past. However, through a contract with the National 

Institute of Building Sciences, FEMA has recently prepared a general loss 

estimation methodology, known as HAZUS, that employs a generalized version of 

this approach. In the HAZUS methodology, building damage functions are based 

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on a standard capacity (pushover) curve for model building type and fragility 

curves that describe the probability of discrete states of damage. Separately, 

building loss functions convert damage to different types of loss including 

casualties, economic losses and loss of function. Damage state probabilities are 

a function of the spectral demand on the structure, determined by the intersection 

of building capacity and earthquake demand spectra. Uncertainty in building 

capacity, damage states, and ground shaking is included in the fragility functions 

that convert spectral demand into damage state probabilities. This approach is 

appealing in that it allows the direct use of the details of an individual building’s 

construction that are important to its earthquake performance, including strength, 

stiffness, and configuration, in the loss estimation process. This approach has 

been adopted for the Detailed Loss Estimation Method found in Appendix B of 

these Recommended Criteria. 

4.2.1 Use of Loss Estimation Methods 

Results from either the Rapid Loss Estimation or Detailed Loss Estimation methods may be 

used to estimate building damage and loss. These data can assist in making economic decisions 

regarding the building, e.g., benefit-cost studies to determine if structural upgrade is warranted. 

Estimates made using the rapid loss estimation method should be considered as representative 

only of average buildings. They should therefore be used with caution since the unique 

structural characteristics of any individual building will affect its vulnerability. While the 

detailed loss estimation method directly takes into account the structural characteristics of a 

building, it also uses general data for other aspects of the loss estimation process including the 

cost of repairing damage of given types, and the replacement value of the building. Hence, 

estimates performed by either of these techniques may require some adjustment by the user to 

better reflect the particular situation. 

Commentary: When applying the rapid loss estimation method to a specific 

building, consideration should be given to such factors as the strength and 

stiffness of the lateral force resisting system, inherent redundancy, physical 

condition, quality of construction, and conformance with building code 

provisions. Buildings having substantial deficiencies would be expected to be 

significantly more vulnerable. Similarly, buildings that have superior earthquake 

resisting characteristics, relative to code requirements, would be expected to be 

less vulnerable. The detailed loss estimation method provides a direct method for 

evaluating these factors. In the rapid method, this can only be accounted for 

qualitatively, using the judgement of the evaluator. 

In addition to these construction characteristics, known to affect building 

performance in earthquakes, a very significant factor that affects the costs 

associated with earthquake damage relates to building occupancy. It is much 

more difficult, and costly, to repair damage in buildings in critical occupancies, 

such as hospitals and semiconductor manufacturing clean rooms, than it is in 

buildings in standard office or residential occupancies. This is both because the 

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finishes and utilities that must be disturbed to conduct structural repairs are more 

complex and expensive, and also, because general working conditions are more 

restrictive. These factors are not directly accounted for by either of the methods. 

4.2.2 Scope of Loss Estimation Methods 

The Rapid Loss Estimation and Detailed Loss Estimation methods may be used to estimate 

direct economic loss related to repair of building damage resulting from the effects of ground 

shaking. These direct losses include costs associated with inspection to determine the extent of 

damage, design and professional services fees, demolition and replacement costs for finished 

surfaces and utilities that must be removed and replaced to allow access for inspection and repair, 

and actual repair construction costs. The methodologies permit estimation of costs related to 

structural repair and to repair of non-structural building features including architectural finishes, 

mechanical and electrical equipment. The methodologies do not include losses related to 

contents including office equipment, inventory, and similar tenant property. 

Ground shaking is the primary, but not the only source of earthquake induced damage, and 

therefore loss that occurs in earthquakes. Other hazards that can result in such losses include 

liquefaction, landsliding, earthquake induced fire and flood. While these hazards typically 

damage only a small percentage of the total inventory of buildings affected by an earthquake, 

they can be far more damaging to those properties that are affected than is ground shaking. 

Regardless, estimates of loss due to these effects are not included in these methodologies. 

In addition to direct economic loss resulting from ground shaking, there are also many other 

types of loss that result from the effects of earthquakes. This includes life loss and injury, as well 

as large economic losses due to interruption of business. Estimation of these losses is also 

beyond the scope of the methodologies presented here. 

4.3 Rapid Loss Estimation Method 

4.3.1 Introduction 

This section presents loss estimation functions that relate seismic demand, resulting primarily 

from ground shaking, to expected loss. The functions are presented in several formats so that 

users can adjust the various loss components to better reflect special knowledge about specific 

buildings. The functions were developed using 1994 Northridge earthquake damage data and 

are, therefore, expected to be representative of steel moment-frame buildings typical of 

California construction prior to 1994. 

In this methodology, losses are quantified in three ways. 

1. Damaged Moment Connections, expressed as a percentage of the total number of moment 

connections in the building. 

2. Connection Restoration Cost, expressed as a percentage of the building replacement value. 

3. Nonstructural Repair Cost, expressed as a percentage of the building replacement value. 

These other repair costs include costs related to restoration of non-structural elements, 

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including fascia, ceilings, and utilities. It does not include costs related to contents such as 

computer systems or stored inventories. 

Commentary: The predictive models for building losses contained in this 

methodology are based on statistical data available from buildings affected by the 

1994 Northridge earthquake. The damage surveys and database used in the 

development of the method dealt with the numbers and types of connection 

damage, and to a lesser extent, with repair costs and nonstructural damage. 

Hence, the primary parameter available and used in the statistical analysis was 

the quantity of damaged moment connections in affected buildings. Reported 

structural repair costs varied widely (some also included costs associated with 

defective welds as opposed to damaged connections), making it impossible to 

derive a reliable direct relationship between seismic demand and connection 

restoration cost. Instead, connection restoration costs were computed for each 

surveyed building as the estimated total number of damaged connections times 

average unit costs for connection repair. For other damage, including 

nonstructural repair costs and other structural repair costs, only very qualitative 

descriptions were reported. Therefore, these other repair costs could not readily 

be ascertained from the Northridge data. The unit costs used in the loss functions 

are provided so that users can adjust loss estimates to better reflect particular 

situations and so that should additional data become available in the future, the 

methodology can be extended in a consistent manner. 

The only structural repair costs directly included in the loss functions 

presented in this methodology are costs related to repair of damaged moment 

resisting connections. Costs related to other structural repairs such as correcting 

permanent interstory drifts are not directly accounted for by these functions. 

However, Section 4.3.4 provides qualitative information that may allow the user 

to develop estimates of the potential additional costs that could be incurred in 

such repairs. 

4.3.2 Seismic Demand Characterization 

Direct damage repair costs are functions of seismic demand resulting primarily from ground 

shaking. The method presented here characterizes seismic demand in three alternative ways. 

1. Modified Mercalli Intensity (MMI) at the building site. MMI is typically derived for a site, 

following an earthquake, based on observation of damage and other earthquake effects at the 

site. Several investigators have developed correlations between observed MMI and estimated 

ground shaking acceleration, velocity and displacement. The MMI values used in these 

Recommended Criteria were derived from estimated peak ground accelerations and velocities 

during the 1994 Northridge earthquake. 

2. Peak Ground Acceleration (PGA) at the building site. This is the geometric mean (square 

root of the product) of the estimated peak values in each of the building’s two principal 

directions. 

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3. Building Pseudo-Drift Ratio (Sd/H). This is defined as the spectral displacement Sd divided 

by the building height H from grade level to main roof. The spectral displacement is the 

geometric mean of the values in each of the building’s two principal directions. The spectral 

displacement is that at the building fundamental period from a site-specific 5% damped 

response spectrum. Consistent units are used so that Sd/H is dimensionless. 

Commentary: Seismic demands are intended to be those caused primarily by 

ground shaking. The Rapid Loss Estimation Method is not intended to cover 

losses governed by other hazards such as ground failure, inundation, and fires 

following earthquakes. 

The damage patterns produced by the 1994 Northridge earthquake exhibited 

considerable scatter. Some buildings reported no connection damage whereas 

others in relatively close proximity had many damaged connections. The reasons 

for this are unclear; however, this random damage pattern has frequently been 

observed in other earthquakes. The scatter may be attributed to a number of 

factors including large uncertainties in the ground motion estimates for each site, 

the effects of individual building configuration and construction quality, and the 

relative thoroughness and accuracy of damage reporting for different buildings. 

Statistical data analysis using numerous different seismic demand measures (e.g., 

MMI, PGA, Peak Ground Velocity (PGV), and Peak Ground Displacement 

(PGD)) as damage predictors did not identify any single parameter as being 

clearly superior for prediction of percentage of damaged connections (FEMA355E). Since no one measure of ground shaking intensity seemed to provide a 

best fit with the available Northridge data, the three measures of ground motion 

intensity presented in these Recommended Criteria were selected based on 

considerations of the probable needs of users. 

MMI was chosen primarily because of its historical use in earlier loss studies 

and the fact that it continues to be used by many practitioners today. MMI is a 

highly subjective parameter intended to be determined after an earthquake, based 

on observed patterns of damage in different areas. It is of course problematic to 

use such an approach to characterize distributions of MMI for a future 

earthquake, that has not yet occurred. A number of researchers have attempted to 

develop correlation functions that relate observed MMI to less subjective 

measures of ground shaking including peak ground acceleration and peak ground 

velocity, which can then be predicted for future earthquakes using various 

attenuation relationships. These predictive models for MMI inherently 

incorporate significant variability and uncertainty. Nevertheless, most 

practitioners who use MMI based approaches to predict losses in future 

earthquakes, first use one of these predictive models for MMI upon which to index 

their loss estimates. 

Consistent with this approach, the MMI values used in the loss functions 

presented here are those inferred from peak ground accelerations and velocities 

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recorded during the 1994 Northridge earthquake, using a predictive model by 

Wald et al. (1998). They are not based on actual damage observations. For a 

given site, there may be considerable difference between the observed MMI and 

the predicted MMI values. 

PGA was chosen because it is an unambiguous, commonly recorded and 

reported, earthquake intensity parameter. One of its shortcomings as a loss 

predictor, however, is that PGA is not reflective of the spectral content of ground 

shaking. Steel moment-frame buildings are typically long-period structures and 

theoretically their response should more closely be related to peak response 

velocity or displacement than to peak ground acceleration. However, these 

quantities are often unavailable for an individual building site, and did not 

provide significantly better correlation with the available data. 

Engineering study of the behavior of steel moment-frame buildings indicates 

that interstory drift is a reasonable parameter for predicting the amount of 

damage experienced by a structure. Therefore, Sd/H was chosen as a ground 

motion intensity index for these Recommended Criteria because it is closely 

related to average interstory drift demands produced in steel moment-frame 

buildings. Also, it includes information about the seismic intensity at the site, and 

the dynamic characteristics of the ground shaking experienced as well as the 

particular building’s dynamic response properties. Unfortunately, statistical 

analysis did not show this to be a better damage predictor than PGA. It is 

believed that the uncertainty in the survey data masks its predictive power. 

Nevertheless, its inclusion here is intended to promote the use of such engineering 

parameters in future loss studies. 

4.3.3 Connection Damage Loss Functions 

Figures 4-1, 4-2 and 4-3 present functions that may be used to estimate Connection Damage 

Ratio (CDR) as a function of Modified Mercalli Intensity (MMI), Peak Ground Acceleration 

(PGA), and Pseudo Interstory Drift Ratio (PIDR), respectively. In these figures, connection 

damage is expressed as the percentage of moment connections within the total number of 

connections in the building’s lateral-force-resisting system in all building directions, that are 

damaged as discussed in Section 2.3. A connection is defined as the attachment of one beam to 

one column. A connection is considered to be either damaged or undamaged (i.e., the relative 

severity of damage is not considered). A connection may be damaged at the beam bottom flange 

location, top flange, or both. Damage may also include the beam web connection and the column 

panel zone. No attempt is made to distinguish between these various types of damage. Defects 

at the roots of the CJP welds between beam and column flanges, which were often categorized as 

damage in buildings affected by the 1994 Northridge earthquake are not considered as damage 

herein. 

Median and 90th percentile loss functions are presented. A set of typical buildings subjected 

to the same seismic demand will exhibit losses over a range. The median loss has the property of 

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having the same numbers of buildings with smaller losses as there are with larger losses. The 

90th percentile loss has the property that 9 out of 10 buildings have losses equal to or lesser in 

magnitude. 

Commentary: Connection damage was the key parameter that was statistically 

evaluated from the 1994 Northridge damage surveys. Connection restoration 

costs (Section 4.3.4) are derived from the connection damage by use of unit repair 

costs. The figures show plots of the actual recorded damage for buildings 

contained in the data set as well as smooth curves that approximately represent 

the median and 90th percentile statistics. The curves were based in part on expert 

judgement that the extent of damage is dependent on seismic demand, even though 

the actual damage data indicates a weak correlation between damage and 

intensity. About 1⁄2 of the buildings in the database experienced no damaged 

connections, and hence many data points are clustered about the horizontal axis 

in the figures. 

The building damage surveys used in the development of the functions 

presented are predominately from buildings covered by the City of Los Angeles 

Ordinance No. 170406 requiring the identification, inspection and repair of 

commercial steel moment-frame buildings subsequent to the 1994 Northridge 

earthquake. The database contained 185 buildings. Implicit in the use of this 

data for loss estimation is the assumption that this sample is representative of 

data for a major metropolitan area. Comparison of the aggregate building 

characteristics (e.g., height and gross area) against census tract data for the 

greater Los Angeles region suggests that the sample is indeed representative of 

the Los Angeles steel moment-frame building population. Whether the sample is 

representative of other metropolitan areas has not been studied. In addition, the 

sample does have certain qualities that are noteworthy. First, residential 

buildings were excluded from the Ordinance and hence are not in the sample. 

Second, most of the seismic demands were in a somewhat limited range (i.e., PGA 

from about 0.25g to 0.45g). Hence, data for PGAs that lie outside this range 

were sparse, and expert opinion was instrumental in defining the loss functions 

there. 

Statistical analysis of the data found that building attributes such as height or 

redundancy (floor area per connection) were not significant parameters affecting 

the percentage of damaged connections. No adjustment factors for these 

characteristics were included herein, nor are they recommended, to adjust the 

estimates made using this data. 

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Steel Moment-Frame Buildings Chapter 4: 

4-9

0

10

20

30

40

50

60

70

80

7 8 9 10

MMI

 % Damaged Connections

90th Percentile

Median

Actual Northridge

Damage Data

Figure 4-1 Connection Damage Ratio vs Modified Mercalli Intensity (MMI)

0

10

20

30

40

50

60

70

80

90

100

0 1 2

PGA (g)

% Damaged Connections

90th Percentile

Median

Actual Northridge

Damage Data

Figure 4-2 Connection Damage Ratio vs Peak Ground Acceleration (PGA)

Loss Estimation
Recommended Seismic Evaluation and Upgrade 

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Chapter 4: Loss Estimation Steel Moment-Frame Buildings 

100 

90 

80 

70 

60 

50 

40 

30 

20 

10 

0 

0 0.01 0.02 0.03 0.04 

90th Percentile 

Median 

Actual Northridge 

Damage Data 

Psuedo Interstory Drift Ratio Sd/H 

Figure 4-3 Connection Damage Ratio vs Building Pseudo Interstory Drift Ratio 

4.3.4 Connection Restoration Cost Functions 

Figures 4-4, 4-5, and 4-6 present connection restoration cost, expressed as a percentage of 

building replacement value, as a function of MMI, PGA and PIDR, respectively. In the 

development of the curves presented in the figures, the average unit cost for connection 

restoration has been taken as $20,000, including costs associated with selective demolition and 

restoration of finishes and utilities to provide access for repair. The building replacement value 

is taken as $125 per sq. ft times the gross building area. 

Commentary: In the development of a typical steel moment-frame building, the 

cost of structural construction is approximately 25% of the total building 

development cost. Thus repair costs on the order of 20% or more approach the 

original cost of constructing the structure. The costs indicated in Figures 4-4, 4-5 

and 4-6 do not include costs associated with repair of damage to elements other 

than moment-resisting connections, for example, column splices, and nonparticipating framing. However, in the 1994 Northridge earthquake, costs of 

these other repairs were not significant. In addition, the above costs do not 

consider the effect of large permanent lateral displacements that can occur in 

damaged frames. Several buildings damaged by the Northridge earthquake 

experienced permanent interstory drifts. Generally, when the permanent drift did 

not exceed a level that was visibly disturbing or interfered with operation of 

elevators, the buildings were not re-plumbed. Re-plumbing buildings that have 

experienced large permanent drifts can be costly, and in many cases may be % Damaged Connections 

4-10
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Steel Moment-Frame Buildings Chapter 4: 

4-11

0

5

10

15

20

25

7 8 9 10

MMI

% Building Replacement Cost

90th Percentile

Median

Estimated Northridge

Repair Costs

Figure 4-4 Connection Restoration Cost vs Modified Mercalli Intensity (MMI)

0

5

10

15

20

25

30

35

0 1 2

PGA (g)

% Building Replacement Cost

90th Percentile

Median

Estimated Northridge

Repair Costs

Figure 4-5 Connection Restoration Cost vs Peak Ground Acceleration (PGA)

Loss Estimation
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Chapter 4: Loss Estimation Steel Moment-Frame Buildings 

30 

25 

20 

15 

10 

5 

0 

0 0.01 0.02 0.03 0.04 

90th Percentile 

Median 

Estimated Northridge 

Repair Costs 

Psuedo Interstory Drift Ratio S

d

/H 

Figure 4-6 Connection Restoration Cost vs Building Pseudo Interstory Drift Ratio 

impractical to accomplish. Thus if a building has experienced large permanent 

interstory drift, the effective cost of structural repair can be larger than indicated 

by these loss functions. 

As a general rule, permanent interstory drift may be on the order of 1/3 to 1/2 

of peak interstory drift. The AISC Standard Practice requires that erection of 

buildings produce a plumb within .005. Permanent interstory drifts of perhaps 

.01 may be tolerable in buildings, while drifts larger than this would probably 

require either straightening or loss of use of the building. These considerations 

have not been accounted for in the above loss functions. 

4.3.5 Nonstructural Repair Cost Functions 

Figures 4-7, 4-8 and 4-9 present nonstructural repair cost, expressed as a percentage of the 

building replacement value, as a function of MMI, PGA and PIDR, respectively. The costs are 

based on HAZUS unit costs and damage states and have been modified by expert opinion 

founded on 1994 Northridge earthquake experience and by engineering judgement. The unit 

costs are taken as Los Angeles commercial office types (professional, technical, and business 

services). Complete repair costs for acceleration-sensitive and drift-sensitive nonstructural 

building components are taken as $42 and $28 per sq. ft, respectively. These unit costs may 

serve as the basis for adjusting the loss functions for particular building situations. % Building Replacement Cost 

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Steel Moment-Frame Buildings Chapter 4: Loss Estimation 

14 

12 

10 

8 

6 

4 

2 

0 

7 7.5 8 8.5 9 9.5 10 

Estimated Northridge Repair 

Costs 

90th Percentile 

Median 

MMI 

Figure 4-7 Nonstructural Repair Cost vs Modified Mercalli Intensity (MMI) 

16 

14 

12 

10 

8 

6 

4 

2 

0 

0 0.5 1 1.5 2 

Estimated Northridge 

Repair Costs 

90th Percentile 

Median 

PGA 

Figure 4-8 Non-Structural Repair Cost vs Peak Ground Acceleration (PGA) % Building Replacement Cost % Building Replacement Cost 

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16 

14 

12 

10 

8 

6 

4 

2 

0 

0 0.005 0.01 0.015 0.02 0.025 0.03 0.035 0.04 

Estimated Northridge Repair 

Costs 

90th Percentile 

Median 

Psuedo Interstory Drift Ratio = Sd/H 

Figure 4-9 Nonstructural Repair Cost vs Building Pseudo Interstory Drift Ratio 

Commentary: Nonstructural repair costs rely heavily on the information from the 

HAZUS project because very sparse quantitative information was available from 

the Northridge damage surveys. Pseudo (or implied) nonstructural repair costs 

were generated for each building in the sample and best-fit curves were generated 

by judgment. The descriptions of nonstructural damage from the Northridge 

building surveys suggested that the repair costs were generally less than that 

indicated by the curves. Hence, the curves were adjusted downward based on 

engineering judgement. % Building Replacement Cost 

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Steel Moment-Frame Buildings Chapter 5: Seismic Upgrade 

5. Seismic Upgrade 

5.1 Scope 

Seismic upgrade measures for steel components and elements of welded steel moment-frame 

(WSMF) structures are described in this chapter. Information needed for simplified and 

systematic upgrade of steel buildings is presented herein. 

5.2 Codes and Standards 

Table 5-1 indicates the general codes, standards, and guideline documents that are applicable 

to seismic upgrades for WSMF structures, and the extent to which they are applicable. 

Table 5-1 Applicable Codes, Standards and Guideline Documents 

Designation Title Applicability 

FEMA-273 NEHRP Guidelines for the Seismic 

Rehabilitation of Buildings 

Provides general performance-based 

guidelines, that are modified herein 

FEMA-302 NEHRP Recommended Provisions for 

Seismic Regulations for New Buildings and 

Other Structures 

Governs the detailing, materials and 

workmanship for new construction 

employed in upgrade design 

AWS D1.1 Structural Welding Code – Steel Governs the requirements for design, 

materials and workmanship for 

structural welding 

AISC/LRFD Specification for the Design of Steel 

Structures 

Provides design requirements for 

bolting, welding, computation of 

member capacities, to the extent 

referenced herein 

NIST/AISC 

(Gross, et al. 

1999) 

Recommendations for Seismic Upgrade of 

Steel Structures 

Provides design procedures for specific 

types of connection upgrades, as 

referenced herein 

Commentary: FEMA-273 provides guidelines for determining force and 

deformation demands for the design of rehabilitation systems for structures to 

meet specific performance objectives. As described in the commentary to Section 

3.1 of this publication, FEMA-273 takes a somewhat different approach to the 

definition of performance objectives than do these Recommended Criteria. Also, 

FEMA-273 was published prior to much of the extensive research on WSMFs 

conducted under this project as well as research conducted by other 

organizations following the 1994 Northridge earthquake. These Recommended 

Criteria contain information that specifically updates the recommendations 

contained in FEMA-273, with regard to the upgrade (rehabilitation) of WSMF 

structures. FEMA-273 provides a more comprehensive treatment on other 

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Chapter 5: Seismic Upgrade Steel Moment-Frame Buildings 

building upgrade issues, including provision of procedures for rehabilitation of 

foundations, diaphragms and nonstructural components. The guidelines 

contained in these Recommended Criteria only address the upgrade of the steel 

frame itself. Refer to FEMA-273 for guidelines on the upgrade of these other 

systems. 

Prior to performing an upgrade on any existing building it is advisable to 

discuss the proposed design criteria with the building official. Although the 

building code for new construction is not intended to apply to existing buildings, 

in some jurisdictions building officials require that upgrades be designed to 

conform to the strength requirements of the current prevailing code, or a fraction 

thereof. In 1991, language was introduced into the Uniform Building Code 

specifically permitting voluntary seismic upgrades of buildings without requiring 

complete conformance with the building code design criteria as long as it could 

be demonstrated that the following conditions did not occur: 

• The upgrade work does not create a structural irregularity or make an existing 

irregular condition more severe 

• The upgrade work does not deliver more load to an existing element than it can 

withstand 

• The upgrade work does not create an unsafe condition. 

Similar language has recently been introduced into the 2000 International 

Building Code. The upgrade criteria contained in these Recommended Criteria 

presume that the above permissive language is incorporated into the local 

building code or that the building official is willing to accept upgrades designed 

to criteria other than that contained in the building code. 

Although these Recommended Criteria suggest that upgrades designed in 

accordance with the criteria need not comply with the strength and drift limits 

specified by the applicable building code for new construction, new work 

performed as part of the upgrade should conform to all materials, detailing, and 

workmanship criteria of the code, as supplemented by these Recommended 

Criteria. 

5.3 Upgrade Objectives and Criteria 

Two approaches are available for seismic upgrade of steel moment-frame structures – a 

Simplified approach and a Systematic approach. In the Simplified approach, modifications are 

made to individual moment-resisting connections to improve their ability to provide ductile 

inelastic behavior. No analyses or evaluations are performed as part of the design of these 

modifications to assess whether the overall structural system is capable of meeting specific 

performance objectives. In the Systematic approach, a complete evaluation of the performance 

capability of the structure is performed in order to verify the performance capability of the 

upgraded structure. Upgrades may include connection modifications, providing supplemental 

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Steel Moment-Frame Buildings Chapter 5: Seismic Upgrade 

lateral force resisting elements, such as braced frames or shear walls, or introducing response 

modification measures such as base isolation or energy dissipation devices. 

Commentary: Throughout the period that steel moment-frame construction has 

been popular, the objective of the building code has been to provide buildings 

with the capability to resist the following: minor earthquakes without damage; 

moderate earthquakes without structural damage but with some nonstructural 

damage; major earthquakes with potentially significant structural and 

nonstructural damage, but not so much damage as to pose a significant threat to 

life safety; and the most severe levels of shaking anticipated to occur at a site 

without collapse. The ability of code-conforming structures actually to provide 

this performance has been mixed. In general, most code-conforming buildings 

have met the latter two goals well, but have experienced more damage at 

moderate levels of shaking than would seem to be desirable. To the extent that 

the code provisions that prevailed at the time a building was designed and 

constructed were adequate to meet these objectives, except that connections were 

more vulnerable to damage than originally believed, the use of the simplified 

upgrade approach, as described in these Recommended Criteria, will restore 

structures to the originally intended performance capability. 

In the simplified upgrade approach, individual moment-resisting connections 

of the structure are upgraded to provide capacity for ductile behavior comparable 

to that presumed to exist at the time of the original design. The adequacy of other 

elements of the structure, including its basic configuration, strength, stiffness, and 

the compactness of sections are not evaluated and are not upgraded. As a result, 

no specific performance can be associated with structures that are upgraded 

using the simplified approach, unless a detailed performance evaluation is 

undertaken, in accordance with the procedures of Chapter 3. 

In the systematic upgrade method a performance evaluation is performed as 

an inherent part of the design evaluation process. This permits upgrade work to 

be designed for specific performance objectives, which may be the same as, 

superior to, or less than those originally intended at the time of building design. 

Regardless of the selected objectives, the systematic approach will provide 

greater confidence in the ability of the structure to actually achieve the intended 

performance than does the simplified approach. 

5.3.1 Simplified Upgrade 

In simplified upgrade, vulnerable connections are upgraded, through a variety of measures, to 

provide more reliable performance of the individual connections. No overall evaluation of the 

performance of the structure, with upgrade modifications, is performed. Presuming that the 

structure, as originally designed and constructed, conformed to the applicable building code 

requirements, but incorporated fracture-vulnerable connections, this method of upgrade could be 

used to restore the structure to its originally intended performance capability. 

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Chapter 5: Seismic Upgrade Steel Moment-Frame Buildings 

In simplified upgrade, the individual beam-column connections of the existing lateral-forceresisting system for the welded steel moment-frame structure are modified to provide equivalent 

interstory drift capacity to that required for a new WSMF structure having the same structural 

system. Existing WSMF structures will typically have been designed, either as Ordinary 

Moment Frames (OMF) or Special Moment Frames (SMF). Chapter 6 of these Recommended 

Criteria provides design criteria for selected pre-qualified connection upgrades, that are accepted 

generically as being capable of providing the necessary drift angle capacity for either OMF 

service, SMF service, or both. Chapter 6 also provides project-specific qualification procedures 

that may be used to affirm that other connection upgrades provide the desired drift angle 

capacity. 

Commentary: The intent of Simplified Upgrade is to reduce the susceptibility of 

moment-resisting beam-column connections detailed and constructed in 

accordance with typical pre-1994 practice to premature, brittle fracture damage. 

When selecting Simplified Upgrade it is inherently accepted that the susceptibility 

of such moment-resisting connections to brittle fracture damage is the only 

significant vulnerability of the structure and that mitigation of this vulnerability 

will result in a structure with acceptable performance characteristics, relative to 

those intended at the time of the original design. This may or may not actually be 

the case, and can be verified only by a detailed performance evaluation. 

Unless original design documents are available, and indicate the design intent 

with regard to the structural system, it should be presumed that the original 

design intent for the structure was to be equivalent to an SMF. If design 

documents are available, these may identify the original intended structural 

system, as being either an SMF, an OMF or a Ductile Moment-Resisting Frame. 

The original design intent for structures indicated as Ductile Moment-Resisting 

Frames should be considered equivalent to that for SMF. 

5.3.2 Systematic Upgrade 

In systematic upgrade, a detailed performance evaluation of the structure is performed in its 

existing configuration and its ability to meet desired performance objectives is determined in 

accordance with the procedures of Chapter 3. If it is found that there is an inadequate level of 

confidence that the structure is capable of meeting the desired performance objectives, then 

structural modifications are performed to improve the probable performance and increase the 

level of confidence. These modifications could include connection improvement measures, such 

as those available for simplified rehabilitation, but could also address systemic issues such as the 

basic strength and stiffness of the structure, the presence of irregularities or other vulnerabilities. 

An iterative process is followed in which a performance evaluation of the building in accordance 

with Chapter 3 is performed assuming proposed modifications are in place, and if the desired 

confidence of achieving the performance objective is not indicated, additional modifications are 

performed. 

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Prior to performing a systematic seismic upgrade, one or more suitable performance 

objectives must be selected as the basis for design. Performance objectives should be selected in 

accordance with Section 3.2. A performance evaluation should be conducted of the structure, to 

determine a level of confidence associated with its ability to meet these performance objectives. 

If sufficient confidence is not attained, then upgrade modifications should be developed, either to 

reduce the response of the structure to earthquake ground shaking, such that acceptable 

confidence of achieving the desired performance is attained, or to increase the capacity of the 

structure to withstand earthquake response and provide acceptable confidence. 

Commentary: Performance objectives, selected in accordance with Section 3.2 

are not completely compatible with those selected in accordance with FEMA-273. 

In FEMA-273, a performance objective is defined as consisting of two parts – a 

desired performance level, of which there are three (Immediate Occupancy, Life 

Safety, and Collapse Prevention) and a desired ground shaking spectrum for 

which this performance level is not to be exceeded. In these guidelines, only two 

performance levels are defined (Immediate Occupancy and Collapse Prevention) 

and a level of confidence with regard to providing the desired performance for a 

given ground shaking hazard is developed. 

The Immediate Occupancy level defined in these Recommended Criteria, is 

similar, but not identical, to the Immediate Occupancy level of FEMA-273. The 

Collapse Prevention level of these Recommended Criteria may be taken as 

equivalent to the Collapse Prevention level of FEMA-273. If it is desired to attain 

performance equivalent to the Life Safety level of FEMA-273, using these 

Recommended Criteria, this may be attained by using 75% of the acceptance 

criteria (e.g., for drift capacities, strength capacities) specified in these guidelines 

for Collapse Prevention. 

To create performance objectives, using these Recommended Criteria, that 

are roughly equivalent to those contained in FEMA-273, it is necessary to 

associate a probability of exceedance, within a specified period (e.g., 50 years) 

with the response spectrum used to define the hazard under the FEMA-273 

criteria. Upgrade designs that provide a 90% confidence level for the desired 

performance level based on global interstory drift, column compression and 

column splice tension and a 50% confidence level for local connection behavior 

at this probability may be deemed equivalent to the intended performance of 

FEMA-273. 

The global interstory drift, capacities and resistance factors contained in 

Chapter 3 are based on typical, regular welded steel moment-frame (WSMF) 

configurations. When adding structural systems that affect the dynamic 

characteristics of the WSMF (e.g., braced frames or shear walls), these default 

factors are no longer valid. For such structural upgrades, the demand and 

resistance factors contained in Chapter 3 may be applied to the calculation of 

confidence relative to local connection, column compression and column splice 

5-5
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Chapter 5: Seismic Upgrade Steel Moment-Frame Buildings 

tension behavior. If the new lateral-force-resisting elements, for example, shear 

walls or braced frames, are designed in accordance with the comparable 

performance objectives of FEMA-273, they may be presumed to provide adequate 

confidence with regard to global building behavior. Alternatively, the detailed 

performance evaluation procedures of Appendix A may be used to confirm global 

behavior. 

5.4 Upgrade Strategies 

A systematic upgrade may be accomplished by any one or more of the following means, as 

required to obtain a structure that provides suitable confidence of capability to provide the 

desired performance: 

• Connection modifications (Section 5.4.1) 

• Removal or lessening of existing irregularities and discontinuities (Section 5.4.2) 

• Global structural stiffening (Section 5.4.3) 

• Global structural strengthening (Section 5.4.4) 

• Mass reduction (Section 5.4.5) 

• Seismic isolation (Section 5.4.6) 

• Supplemental energy dissipation (section 5.4.7) 

Commentary: A building’s response to earthquake ground shaking results in the 

development of forces and deformations in the structure. In Chapter 3 of these 

Recommended Criteria, a procedure is defined for determining a level of 

confidence with regard to the ability of a structure to resist these forces and 

deformations with a defined probability of exceeding one or more performance 

levels. This confidence level is tied to the confidence parameter l calculated as 

the ratio of the factored demands g gaD to the factored capacity fC to resist these 

demands. Values of the parameter l less than 1 indicate relatively high 

confidence, while values above 1 indicate progressively lower confidence. 

If upon evaluation in accordance with Chapter 3, it is found that an 

inadequate level of confidence is obtained with regard to the ability of the 

structure to meet a desired performance objective, an upgrade can be performed 

to improve this confidence. To be effective, this upgrade must be able either to 

increase the capacity of the structure, and its various elements to resist the forces 

and displacements induced by earthquake response, or alternatively, the amount 

of force and deformation that a structure develops (the demands) must be 

reduced. As a third alternative, it may be possible to attain a higher level of 

confidence with regard to the probable performance of a structure by obtaining 

better information on the structure’s construction and by performing more 

detailed and certain analyses of the structure’s response to ground shaking. The 

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following sections provide information on alternative methods of modifying a 

structure to either increase its capacity or decrease the demands. Appendix A 

provides guidelines for improving confidence with regard to the structure’s 

performance, through the use of more accurate analyses and evaluations. 

5.4.1 Connection Modifications 

Connection modifications are intended to upgrade the ability of individual connections to 

withstand expected rotational deformations with suitably low probability of unacceptable 

damage. This is judged to have been achieved when the ratio of factored drift angle capacity fC 

of the individual connections to withstand the factored demands g gaD determined from an 

analytical evaluation of structural performance, results in an acceptable confidence index l. 

Connection upgrades accomplish this in two ways. First, the upgrades directly improve the 

interstory drift angle capacity of individual connections, resulting in a reduced value of l for 

local behavior. Second, if connections are upgraded to Special-Moment-Frame-compatible 

detailing, the connections are converted from type 2 (brittle behavior) to type 1 (ductile behavior) 

permitting use of increased global interstory drift capacities and reduced demand factors. Chapter 

3 provides more information on these issues. Chapter 6 presents a series of pre-qualified 

connection upgrades, together with design procedures, the limiting parameters for which these 

upgrades are pre-qualified, and the drift angle capacities of the upgraded connections. Chapter 6 

also presents a project-specific connection qualification procedure for use in determining 

appropriate drift angle capacities and capacity factors, for connection upgrades that are not 

included in the prequalifications. 

Commentary: Connection upgrades are a method of increasing the local capacity 

of the individual connections to withstand inelastic deformation demands, as 

measured by drift angle. These upgrades do not, in general, reduce the demands 

produced in a structure by earthquake response. Therefore, connection upgrades 

are not, by themselves, particularly effective in improving the performance of 

structures that experience excessive demands due to inadequate frame stiffness or 

strength, or inappropriate frame configuration. Such vulnerabilities are better 

addressed with other upgrade strategies. For many structures, it may be 

necessary both to reduce the demands produced by earthquake response as well 

as increase the capacity of the individual connections to resist this response. In 

such cases, connection upgrades should be performed together with other 

upgrade strategies. 

Although connection upgrade strategies directly address the single most 

common vulnerability of steel moment-frame structures – connections prone to 

premature brittle fracture – these upgrades can be quite costly, particularly in 

large structures with many connections. In some cases, it may be more cost 

effective to adopt strategies intended to reduce demands on connections rather 

than to increase individual connection capacities. 

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Some connection upgrade details have the potential to grossly affect the 

inelastic response behavior of frames. For example, some connection upgrades 

may shift the zones of plastic deformation from the beam column-joint to the 

beam, column or panel zone. Such modifications of inelastic response behavior 

will alter the demands placed on the individual connections, as well as the frame 

as a whole, and should be considered when connection upgrade strategies are 

adopted. 

Connection upgrades that improve the drift angle capacity of the connections 

compatible with Special Moment-Frame requirements for new construction also 

result in a decrease in uncertainty relative to probable frame behavior. This is 

because of the reduced propensity for brittle fracture of the connections. This 

reduction in uncertainty is reflected in the use of demand factors appropriate for 

Type 1 connections, as described in Chapter 3. 

5.4.2 Lessening or Removal of Irregularities 

Many existing welded steel moment-frame buildings incorporate one or more structural 

irregularities. Some irregularities, such as soft stories, weak stories, torsional irregularities, and 

discontinuous structural systems can result in poor structural performance. Typically this poor 

performance occurs due to the concentration of force and inelastic deformation demand in the 

area of the irregularity. Often, the structural elements in the area of the irregularity are incapable 

of withstanding these locally increased demands. Structural upgrades that remove or lessen these 

irregularities have the effect of decreasing this concentrated demand resulting in a more uniform 

distribution of deformation and energy dissipation throughout the structure. 

A structural irregularity should not be considered to be a problem unless a structural 

performance evaluation, conducted in accordance with Chapter 3 of these Recommended 

Criteria, indicates that structural demands, e.g., interstory drift or column axial load, in the area 

of the irregularity are in excess of the acceptance criteria for the desired structural performance 

level. Where an undesirable irregularity exists, it can usually be eliminated or reduced through 

the local introduction of new structural elements or through strengthening and stiffening of 

existing elements. When such features are introduced, a re-evaluation of the entire structure 

should be performed to ensure that the measure will result in adequate performance and that 

some new irregularity or vulnerability has not been inadvertently introduced into the structure. 

5.4.3 Global Structural Stiffening 

Damage to both structural and nonstructural elements is closely related to the amount of 

deformation induced in a building by its response to ground shaking. Global structural stiffening 

is intended to directly reduce the amount of this lateral deformation through introduction of 

stiffening elements. Although reinforcement of connections often results in some structural 

stiffening, this is typically not a significant effect and is not by itself adequate to result in 

substantial reductions in lateral deformation. In order to have a noticeable effect on performance, 

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Steel Moment-Frame Buildings Chapter 5: Seismic Upgrade 

substantial stiffening is typically required. In some cases it may be possible to accomplish this 

by converting some beam-column connections that were not originally connected for momentresistance, into moment-resisting connections. If this is done, care must be taken to ensure that 

the beams and columns are adequate for the stresses induced by this approach. The most 

effective way to increase the stiffness of a WSMF structure is to add braced frames and/or shear 

walls to the seismic force resisting system. 

Although global stiffening is effective in reducing the amount of deformation induced in a 

structure due to its earthquake response, it also typically results in some increase in the level of 

forces delivered to the structure and its nonstructural components. When evaluating the 

performance of the upgraded structure, it is important to evaluate all elements, including those 

that were determined to be adequate prior to the upgrade, as the additional forces delivered to 

these elements by the stiffened structure may result in poorer performance than previously 

indicated in evaluations of the performance of the existing structure, without such upgrades. 

FEMA-273 provides modeling guidance and acceptance criteria for bracing and shear wall 

elements used to structurally stiffen a steel moment-frame structure. Upgrades using this strategy 

shall be conducted by designing the upgrade elements using the guidelines of FEMA-273. The 

performance of WSMF elements of the structure, including connections, columns and column 

splices shall be evaluated using the procedures of Chapter 3. If the new stiffening elements have 

been designed in accordance with the guidelines of FEMA-273, it may be presumed that a 90% 

level of confidence with regard to global building behavior can be attained. If desired, the user 

may confirm the adequacy of global performance of the upgraded structure, using the procedures 

of Appendix A of this document to determine the global drift capacity. 

5.4.4 Global Structural Strengthening 

Typically, WSMF structures do not exhibit poor performance as a result of inadequate 

strength to resist lateral forces. Rather, they exhibit poor performance because they are 

excessively flexible, have excessive irregularities or have vulnerable details and connections. 

However, if a performance evaluation of a WSMF structure indicates inadequate performance 

due to a global lack of adequate ability to resist lateral forces, such as those produced by ground 

shaking, strengthening of the structure can be achieved by many of the same means used for 

structural stiffening, as indicated in Section 5.4.3. In addition, global strengthening can be 

achieved by cover plating members of the lateral-force-resisting system in order to provide them 

with additional strength. When global strengthening is performed, the building, including 

structural and nonstructural elements, is likely to experience greater forces. Therefore, when 

evaluating the performance of the upgraded structure, it is important to evaluate all elements, 

including those that were determined to be adequate prior to the upgrade, as the additional forces 

delivered to these elements by the stiffened structure may result in poorer performance than 

previously indicated in evaluations of the performance of the existing structure, without such 

upgrades. 

FEMA-273 provides modeling guidance and acceptance criteria for bracing and shear wall 

elements used to structurally stiffen or strengthen a WSMF structure. Upgrades using this 

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strategy shall be conducted by designing the upgrade elements using the guidelines of FEMA273. The performance of elements of the structure, including connections, columns and column 

splices shall be evaluated using the procedures of Chapter 3. If the new strengthening elements 

have been designed in accordance with the guidelines of FEMA-273, it may be presumed that a 

90% level of confidence with regard to global building behavior can be attained. If desired, the 

user may confirm the adequacy of global performance of the upgraded structure, using the 

procedures of Appendix A to determine the global drift capacity. 

Commentary: Since WSMF structures are anticipated to exhibit significant 

response within the inelastic range, it can be difficult to determine if the inability 

of a structure to provide adequate performance is a result of inadequate strength 

as opposed to stiffness. Generally, global structural strength is closely related to 

a structure’s ability to provide Immediate Occupancy performance, while global 

stiffness is more closely related to Collapse Prevention performance. An inability 

of a structure to provide adequate confidence of achievement of Collapse 

Prevention performance will usually be most effectively mitigated through 

addition of structural stiffness, rather than strength. Similarly, an inability of a 

structure to provide adequate confidence of achievement of Immediate Occupancy 

performance can often best be addressed through addition of global structural 

strengthening. 

5.4.5 Mass Reduction 

The reduction of mass in a structure can improve its performance in several ways. One effect 

of mass reduction is a decrease in the periods of vibration of the structure. Since buildings of 

decreased period generally exhibit lower deformation response than do buildings of longer 

period, this results in decreased deformation and damage. The seismic forces experienced by a 

structure are proportional to the acceleration induced by the earthquake and the structure’s mass. 

By reducing the structure's mass it is possible to reduce directly the amount of seismic force 

induced in the structure, which also reduces the potential damage. 

Methods of reducing the mass of a steel moment-frame structure can include: replacement of 

heavy exterior cladding systems with lighter systems; removal of unused equipment and storage 

loads; replacement of masonry partition walls with lighter systems; and removal of one or more 

stories. As with other upgrade techniques, a complete re-evaluation of the upgraded structure's 

performance should be conducted, following development of an upgrade alternative. 

Commentary: The most beneficial effect of mass reduction as an upgrade 

strategy is that it leads to a shortening of the structural period, and a 

corresponding reduction in the spectral displacement demand on the structure, 

produced by typical earthquake ground motions. However, period is related to 

mass through a square root relationship. Thus, substantial reductions in mass 

are necessary to have a meaningful effect on lateral displacement demand. 

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5.4.6 Seismic Isolation 

Seismic isolation is a relatively new method of improving the seismic performance of an 

existing structure. Seismic isolation improves structural performance through two basic effects. 

First, it is used to significantly lengthen the period of the structure, potentially in combination 

with the introduction of significant damping. The combined effect of the change in the 

structure’s period and the introduction of supplemental damping results in greatly reduced 

seismic inertial forces on the building. Isolation systems are also typically designed such that 

they are more flexible than the supported structure, such that most of the earthquake induced 

deformation and energy dissipation is accommodated within the isolation system, rather than 

being transmitted to the structure. The result is that the components of the isolation system 

experience very large deformation and energy dissipation demands, while the structure above the 

isolation system sees relatively low levels of seismic induced lateral forces and deformations, and 

therefore, low levels of damage. 

Seismic isolation tends to be most effective as an upgrade measure when a relatively heavy 

and stiff superstructure is mounted on relatively flexible isolators. Typically the period of the 

isolated structure (including the isolation system) is on the order of 2 to 3 seconds. Isolation is 

most effective when the initial period of the non-isolated structure is on the order of 1 second or 

less. Since most steel moment-frame (WSMF) structures have periods in excess of 1 second, this 

will not often be an effective method of upgrading WSMF structures, unless it is combined with 

supplemental global stiffening of the structure. 

FEMA-273 provides modeling guidelines and acceptance criteria for isolation systems for use 

in performance evaluation of isolated structures. Upgrades using this strategy shall be conducted 

by designing the upgrade elements using the guidelines of FEMA-273. The performance of 

elements of the structure shall then be evaluated using the procedures of Chapter 3, with the 

mathematical model modified to include the effects of the upgrade elements on structural 

response. For purposes of performance evaluation, the interstory drift of the isolation system 

shall be neglected. Global interstory drift demand shall be taken as the maximum of the 

interstory drifts predicted for the superstructure, considering the effects of the isolation system in 

the model. 

Commentary: Performance evaluation conducted in accordance with the 

procedures of Chapter 3 uses maximum predicted interstory drift demand as one 

of the primary parameters evaluated. The primary effect of base isolation is to 

substantially reduce the interstory drift demand within the structure. The base 

isolation system should be designed in accordance with the procedures of FEMA273. The performance of the superstructure should be evaluated using the 

procedures of Chapter 3 and taking the interstory drift demand as that predicted 

for the frame, in an analysis in which the base isolation system as well as the 

frame is modeled. 

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5.4.7 Supplemental Energy Dissipation 

The intent of seismic upgrades employing supplemental energy dissipation devices, also 

called dampers, is to reduce the amount of deformation induced in the structure during its 

response to ground shaking. In this respect, it is similar to upgrades accomplished through global 

structural stiffening. However, rather than introducing stiffening to a structure, this upgrade 

technique reduces deformation through the dissipation of energy within a series of devices that 

are introduced into the structure as part of the upgrade. The effect of this dissipated energy is to 

increase the structure’s effective damping, and thereby, to reduce its lateral displacement 

response. 

A number of different types of energy dissipation devices are commercially available. These 

include fluid-viscous dampers, visco-elastic dampers, friction dampers, and hysteretic dampers. 

Each of these devices has unique force-displacement-velocity relationships, and therefore affect 

the structure’s response in a somewhat different manner. 

The energy dissipated by a damping device is the integrated product of the amount of force 

the device exerts on the structure (or is exerted on the device by the structure) and the distance 

through which this force acts. In many ways, welded steel moment-frame structures are ideal 

candidates for upgrades employing energy dissipation devices because they are inherently 

flexible structures permitting damper elements to dissipate large amounts of energy at relatively 

low force levels. This is important because large damper forces can create large concentrated 

forces in the structure. 

Energy dissipation devices are typically introduced into a structure as part of a braced frame, 

where the devices are either introduced in series with the braces in the frame, or actually serve as 

the braces in the frame. Upgrades using this strategy should be conducted by designing the 

upgrade elements using the guidelines of FEMA-273. The performance of elements of the 

structure should then be evaluated using the procedures of Chapter 3, with the mathematical 

model modified to include the effects of the upgrade elements on structural response. 

5.5 As-Built Conditions 

5.5.1 General 

Prior to performing an upgrade design, sufficient information on the configuration and 

material properties of the existing structure must be obtained to permit a detailed evaluation, in 

accordance with Chapter 3. Refer to Chapter 2 for criteria on obtaining as-built information. 

Quantification of in-place material properties and verification of the existing system 

configuration and condition are necessary to analyze or evaluate a building. Chapters 2 and 3 

identify properties requiring consideration and provide criteria for their acquisition. Condition 

assessment is an important aspect of planning and executing seismic upgrade of an existing 

building. One of the most important steps in condition assessment is a visit to the building for 

visual inspection. 

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The extent of in-place materials testing and condition assessment that must be accomplished 

is related to availability and accuracy of construction and as-built records, the quality of materials 

used and construction performed, and the physical condition of the structure. Data such as the 

properties and grades of material used in component and connection fabrication may be 

effectively used to reduce the amount of in-place testing required. The design professional is 

encouraged to research and acquire all available records from original construction. 

5.5.2 Material and Section Properties 

Material and section properties of existing components shall be determined in accordance 

with the criteria outlined in Chapter 2. 

5.6 Upgrade Components 

New components, constructed as part of upgrades of existing WSMF structures shall conform 

to the requirements of this section. 

5.6.1 Material Specifications 

Structural steel should conform to the specifications and grades permitted by the building 

code, unless a project-specific qualification testing program is performed to demonstrate 

acceptable performance of alternative materials. 

5.6.2 Material Strength Properties 

The AISC Seismic Provisions (AISC, 1997) state: 

When required by these provisions, the required strength of a connection or related 

member shall be determined from the Expected Yield Strength Fye of the connected 

member, where 

Fye = RyFy (5-1) 

The Provisions further state that “Ry shall be taken as 1.5 for ASTM A36 and 1.3 for A572 

Grade 42. For rolled shapes and bars of other grades of steel and for plates, Ry shall be taken 

as 1.1. Other values of Ry are permitted to be used if the value of Fye is determined by testing 

that is conducted in accordance with the requirements for the specified grade of steel.” 

ASTM has recently issued a new specification, A992, for structural steel shape. This 

specification is similar to the ASTM A572 specification for Grade 50 steels, except that more 

restrictive limits apply to the permissible variation in yield strength, the ratio of yield to tensile 

strength and certain other properties, than contained in ASTM A572. This material specification 

was specifically developed by the steel industry in response to concerns raised by structural 

engineers with regard to the large variations in properties inherent in the A572 specification, and 

the difficulties this presented with regard to design for inelastic behavior and seismic resistance. 

The A992 material will eventually become the recommended basic grade of steel for use in 

seismic force resisting systems. Since material has only recently been produced under this 

specification, statistical data on the actual variation of strength properties produced by the mills 

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is not currently available. Until such data does become available, use of the Ry values indicated 

for ASTM A572, Grade 50 is recommended. 

5.6.3 Mathematical Modeling 

The stiffness and strength of upgrade elements shall be included in the mathematical model 

using the same criteria provided for modeling of existing elements as outlined in Chapter 3. 

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Steel Moment-Frame Buildings Chapter 6: Connection Qualification 

6. Connection Qualification 

6.1 Scope 

This chapter provides performance qualification data for various types of connections, 

together with criteria for analysis and design of connections for the upgrade of existing steel 

moment-frame (WSMF) structures. Included herein are general criteria that are generic to most 

connection upgrade types, and recommendations for specific connection upgrade details of 

connections intended to be prequalified for use in seismic upgrades. Each of the connection 

prequalifications is limited to specific conditions for which they are applicable, including 

member size ranges, grades of material and other details of the connection. Also included in this 

chapter are procedures for qualification of connections and connection upgrades, which have not 

been prequalified or are proposed for use outside the limits of their prequalification as set forth 

herein. 

Commentary: The 1988 Uniform Building Code (ICBO, 1988) introduced a single 

pre-qualified moment connection design, representative of prevailing west coast 

practice at the time. The “qualification” of this connection was based primarily 

on the research of Popov and Stephen in the early 1970’s, and the belief that this 

connection was capable of providing acceptable strength and ductility for service 

in all frames that otherwise met the provisions of the building code. The UBC 

pre-qualified connection was subsequently adopted into the 1992 AISC Seismic 

Provisions and then into model codes nationwide. Although the building codes 

did not formally adopt the pre-qualification of this standard connection until the 

late 1980s and early 1990s, this connection detail had seen widespread use in 

WSMF construction since the 1970s. 

The discovery of many fractures in buildings incorporating this standard 

detail, following the Northridge earthquake, demonstrated the ineffectiveness of 

the pre-qualified connection as it was being used in modern practice. Subsequent 

research conducted under this project, and by others, has demonstrated that many 

types of connections that have the strength to develop the plastic moment capacity 

of the connected elements, do not have the capability to do so in a ductile manner 

over repeated cycles of loading. Further, this research has shown that inelastic 

deformation demands in some frame structures can be significantly larger than 

those that have historically been presumed as the basis for the codes. 

Following the 1994 Northridge earthquake, the pre-qualified connection 

contained in the building code was deleted by means of an emergency code 

change. In its place, a provision was substituted requiring that the designer 

demonstrate that whatever connection was used is capable of sustaining the 

necessary inelastic deformation demands. Qualification of this capacity was by 

prototype testing. In the time since, a significant number of connection 

assemblies have been tested, allowing new prequalifications to be developed. 

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Those prequalifications that are applicable to the upgrade of existing structures 

appear in this document. 

Although a number of prequalified connection upgrades are available, it is 

conceivable that designers may wish to utilize other connection upgrade designs 

or to use a pre-qualified design under conditions that are outside those for which 

they have been prequalified. In these cases, a project-specific, qualification-bytest procedure is still required. The requirements for such a qualification 

procedure are also given in this chapter. 

Finally, this chapter presents qualification and modeling data needed for the 

assessment of performance of the typical pre-Northridge style connection and of 

various types of simple gravity connections, for use in performance evaluation of 

existing structures. 

6.2 Performance Data for Existing Connections 

This section provides modeling criteria and performance data for use in assessing the 

performance of existing moment-resisting and simple connections typically found in existing 

welded steel moment-frame buildings. These connections are not prequalified for use in the 

lateral-force-resisting systems of new structures. For each connection type, the following 

quantities are defined: 

qSD = median total connection drift angle at which strength degradation occurs, radians. 

For existing brittle connections, this corresponds to the median estimate of drift 

angle at which brittle fracture initiates 

qIO = median drift angle capacity for Immediate Occupancy performance, radians 

qU = median drift angle at which connection looses gravity load carrying ability, used 

as the limit state for Collapse Prevention performance 

f = a resistance factor applied to qIO, or qU, as appropriate 

6.2.1 Welded Unreinforced Fully Restrained Connection 

The data contained in this section applies to performance evaluation of existing buildings 

with the typical welded, unreinforced, moment-resisting connection, commonly present in 

WSMF buildings constructed prior to the 1994 Northridge earthquake. Figure 6-1 presents a 

detail for this connection. It is characterized by rolled wide flange beams connected to the strong 

axis of wide flange column sections, with the connection of the beam flanges to column flange 

through complete joint penetration (CJP) groove welds. Welding has typically been performed 

using the Flux Cored Arc Welding process and with weld filler metals without specific rated 

notch toughness. Weld backing and weld tabs are commonly left in place. Beam webs are 

connected to the column with a single plate shear tab, welded to the column and bolted to the 

beam web. In some forms of the connection, there are supplemental welds of the shear tab to the 

beam. Doubler plates, reinforcing the shear capacity of the column panel zone, and beam flange 

continuity plates at the top and bottom of the panel zone may or may not be present. 

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Plate 

Plate 

Beam 

Column 

Beam Flange 

Continuity 

Doubler 

CJP 

Figure 6-1 Welded Unreinforced Fully Restrained Connection (pre-1994) 

Commentary: The data presented in this section is not specifically applicable to 

forms of this connection that employ weld metals with significant notch 

toughness. Some older buildings, particularly those erected prior to about 1964, 

may have welds deposited by the Shielded Metal Arc Welding (SMAW) process. 

Some such welds may have significant notch toughness, on the order of 40 ft-lbs 

at normal service temperatures. Limited testing of such connections indicates 

that they may have better inelastic deformation capacity than do connections 

employing weld material with lower notch toughness. Refer to Section 6.6.1 for 

data on connections with notch-tough weld metal. 

The performance data provided in this section also is not specifically 

applicable to forms of the connection in which the beam web is directly welded to 

the column flange. Limited testing of such connections indicates that they are 

capable of providing somewhat better inelastic deformation capacity than similar 

connections with bolted beam webs. However, there are not sufficient data 

available to permit separate performance qualification of this connection type. 

The performance data provided herein may be conservatively applied to that 

connection type, or alternatively, project-specific qualification testing of such 

connections may be performed. 

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The connection performance data contained herein has been based on testing 

of connection assemblies in which the beams are connected to the major axis of 

the column. Connections in which beams are connected to the minor axis of 

columns are known to have similar, and perhaps, more severe vulnerability than 

major axis connections. However, insufficient data are available to permit 

quantification of this performance. Connections employing box columns are 

beyond the scope of this section. 

6.2.1.1 Modeling Assumptions 

6.2.1.1.1 Linear Analysis 

Elastic analysis models of structures with Welded Unreinforced Fully Restrained 

Connections should be based on the assumption that the connection provides a fully rigid 

interconnection between the beam and column, located at the centerline of the column. 

Alternatively, realistic assumptions with regard to panel zone flexibility may be made, as 

indicated in Section 3.5.2.2. 

6.2.1.1.2 Nonlinear Analysis 

Nonlinear analysis models of structures with Welded Unreinforced Fully Restrained 

Connections should be based on the assumption that the connection provides a fully rigid 

interconnection between the beam and column, located at the centerline of the column, until the 

connection panel zone, the beam or the column yields, or a total interstory drift angle qSD, from 

Table 6-1 is reached. The expected yield strength of the material, as indicated in Section 2.5 

should be used to calculate the yield capacity of beams, columns, and panel zones. If yielding 

occurs at total interstory drift angles less than qSD, the yielding element should be assumed to 

exhibit plastic behavior. At interstory drifts greater than qSD the connection should be assumed 

to be capable of transmitting 20% of the expected plastic moment capacity of the girder until a 

total interstory drift angle qU, obtained from Table 6-1, occurs. At interstory drift angles greater 

than qU, the connection should be presumed to have negligible strength. 

6.2.1.2 Performance Qualification Data 

Table 6-1 presents the applicable performance qualification data for welded unreinforced 

fully restrained moment-resisting connections, conforming to typical practice prior to the 

Northridge earthquake. 

6.2.2 Simple Shear Tab Connections – with Slabs 

The data contained in this section applies to the typical single plate shear tab connection 

commonly used to connect beams to columns for gravity loads, when moment-resistance is not 

required by the design, and when concrete slabs are present. Figure 6-2 presents a detail for this 

connection. It is characterized by rolled wide flange beams connected to either the major or 

minor axis of wide flange column sections. Beam webs are connected to the column with a 

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single plate shear tab, welded to the column and bolted to the beam web. A concrete floor slab, 

or slab on metal deck is present at the top flange of the beam. 

Table 6-1 Performance Qualification Data – 

Welded Fully Restrained Connection (pre-1994) 

Data Applicability Limits 

Hinge location distance sh At distance db/3 from face of column, unless shear strength of panel 

zone is less than shear corresponding to development of the flexural 

strength of beams at the connection, in which case, the hinge 

location should be taken at the column centerline. 

Maximum beam size Unlimited 

Beam material A36, A572, Gr. 50 

Maximum column size Unlimited 

Column steel grades A36, A572, Gr. 50 

Performance Data 

Strength degradation rotation - qSD, radians 0.061-0.0013db 

Immediate Occupancy rotation - qIO, radians 0.01 radian, but not greater than qSD 

Resistance factor, Immediate Occupancy, f 0.8 

Collapse Prevention drift angle - qU – radians 0.053-0.0006db 

Resistance factor, Collapse Prevention, f 0.8 

Notes: db= beam depth, inches 

Commentary: Although shear tab connections of the type shown in Figure 6-2 

are not typically included in design calculations as part of the lateral-forceresisting system, research conducted in support of these Recommended Criteria 

(FEMA-355D) indicates that they are capable of providing both non-negligible 

strength and stiffness. Since the typical steel moment-frame structure will have 

many such connections, the presence of these connections converts the gravity 

load framing into a highly redundant reserve system to provide additional 

stiffness and strength for the building after the primary system comprised of fully 

restrained connected framing has been damaged. 

When these connections are loaded such that the top beam flange acts in 

compression, the slab can act compositely with the beam. When this behavior 

occurs, the slab will bear against the column and significant moments can 

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develop through a couple consisting of the slab in compression and the shear tab 

in tension. This behavior is limited by local crushing of the slab in compression, 

which behavior initiates at moderate interstory drift angles. Following crushing 

of the slab, the connections acts as if the slab were not present, and provides 

relatively modest flexural resistance until very large rotations. Ultimately, at very 

large rotations, the beam compressive flange will bear against the column, again 

resulting in development of large moments. Since the beam flange does not crush, 

this typically results in failure of the shear tab, in tension. 

The criteria for modeling these connections, presented here, neglects the 

effect of the slab as described above. This is because this behavior occurs only 

for one direction of loading, and also, because at large deformations, this 

behavior degrades. However, nothing in this document would preclude more 

accurate modeling of these connections, that accounts for the slab effects. 

FEMA-355D provides information that may be useful for this more complex 

modeling. 

Major Axis of Column Minor Axis of Column 

Figure 6-2 Typical Simple Shear Tab Connection with Slab 

6.2.2.1 Modeling Assumptions 

When included in the analytical model used to predict earthquake induced demands, the 

stiffness and hysteretic characteristics of framing with simple shear tab connections should be 

taken in accordance with the recommendations of this section. 

6.2.2.1.1 Linear Analysis 

The connection stiffness should be explicitly modeled as a rotational spring that connects the 

beam to the column. The spring stiffness, Kq should be taken as: 

Kq = 28000(dbg - 5.6) (6-1) 

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where dbg is the depth of the bolt group in inches and Kq is in units of k-inches per radian. In lieu 

of explicit modeling of the connection, beams that frame into columns with simple shear tab 

connections may be modeled with an equivalent rigidity, EIeq taken as: 

1 EI eq = 

6h 1 (6-2) 

+ 2

lb Kq EI b 

where:

E = the modulus of elasticity, kip/square inch

h = the average story height of the columns above and below the beam, inches

Ib = the moment of inertia of the beam, (inches)

4

lb = the beam span center to center of columns, inches

6.2.2.1.2 Nonlinear Analysis 

The connection should be explicitly modeled as an elastic-perfectly-plastic rotational spring. 

The elastic stiffness of the spring should be taken as given by Equation 6-1. The plastic strength 

of the spring should be determined as the expected plastic moment capacity of the bolt group, 

calculated as the sum of the expected yield strength of the bolts and their distance from the 

neutral axis of the bolt group. 

6.2.2.2 Performance Qualification Data 

Table 6-2 presents the applicable performance qualification data for shear tab connections of 

beams to columns, with slabs present. 

6.2.3 Simple Shear Tab Connections – Without Slabs 

The data contained in this section applies to the typical single plate shear tab connection 

commonly used to connect beams to columns for gravity loads, when moment-resistance is not 

required by the design and slabs are not present. Figure 6-3 presents a detail for this connection. 

It is characterized by rolled wide flange beams connected to either the major or minor axis of 

wide flange column sections. Beam webs are connected to the column with a single plate shear 

tab, welded to the column and bolted to the beam web. Diaphragms may not be present, and if 

present consist of wood sheathing, unfilled metal deck, or horizontal steel bracing. 

Commentary: Shear tab connections without slabs present behave in a very 

similar manner to shear tabs with slabs, except that the composite behavior with 

the slab discussed in the previous section does not occur. Since the modeling 

criteria for connections with slabs neglect the strength contribution of the slab, 

the criteria presented herein for connections without slabs are essentially 

identical to those presented in the previous section. 

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Table 6-2 Performance Qualification Data – Shear Tab Connections with Slabs 

Data Applicability Limits 

Hinge location distance sh at center line of bolts 

Maximum beam size Unlimited 

Beam material A36, A572, Gr. 50 

Maximum column size Unlimited 

Column steel grades A36, A572, Gr. 50 

Performance Data 

Strength degradation rotation - qSD, radians 0.039-0.0002dbg 

Immediate Occupancy rotation - qIO, radians 0.025, but not greater than qSD 

Resistance factor, Immediate Occupancy, f 0.90 

Collapse Prevention drift angle - qU – radians 0.16-0.0036dbg 

Resistance factor, Collapse Prevention, f 0.80 

Note: dbg = bolt group depth, measured from center of top bolt to center of bottom bolt, inches 

Major Axis of Column Minor Axis of Column 

Figure 6-3 Typical Simple Shear Tab Connection Without Slab 

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6.2.3.1 Modeling Assumptions 

Shear tab connections without slabs present should be modeled the same as shear tab 

connections with slabs present, as indicated in Section 6.2.2.1, except that for nonlinear analysis, 

performance qualification data shall be as indicated in Table 6-3. 

6.2.3.2 Performance Qualification Data 

Table 6-3 presents the applicable performance qualification data for shear tab connections of 

beams to columns, without slabs present. 

Table 6-3 Performance Qualification Data – Shear Tab Connections (No Slab) 

Data Applicability Limits 

Hinge location distance sh At center line of column 

Maximum beam size Unlimited 

Beam material A36, A572, Gr. 50 

Maximum column size Unlimited 

Column steel grades A36, A572, Gr. 50 

Performance Data 

Strength degradation rotation - qSD, radians 0.16-0.0036dbg 

Immediate Occupancy rotation - qIO, radians 0.030, but not greater than qSD 

Resistance factor, Immediate Occupancy, f 0.90 

Collapse Prevention drift angle - qU – radians 0.16-0.0036dbg 

Resistance factor, Collapse Prevention, f 0.80 

Note: dbg = bolt group depth, measured from center of top bolt to center of bottom bolt, inches 

6.3 Basic Design Approach for Connection Upgrades 

This section provides recommended criteria on basic principles of connection upgrade 

design, including selection of an appropriate connection upgrade detail, estimation of locations of 

inelastic behavior (formation of plastic hinges), determination of probable plastic moment at 

hinges, determination of shear at the plastic hinge, and determination of design strength demands 

at critical sections of the assembly. The designer should utilize these basic principles in the 

calculations for all connection upgrades, unless specifically noted otherwise in these 

Recommended Criteria. 

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6.3.1 Frame Configuration 

Upgraded frames should be proportioned and detailed so that the required drift angle of the 

frame can be accommodated through elastic deformation and the development of plastic hinges 

at pre-determined locations within the frame. Figure 6-4 indicates a frame in which inelastic 

drift is accommodated through the development of plastic flexural deformation (plastic hinges) 

within the beam span, remote from the face of the column. Such behavior may be obtained by 

locally stiffening and strengthening type FR connections, using cover plates, haunches and 

similar detailing, such that the ratio of flexural demand to plastic section capacity is maximum at 

these interior span locations. Other locations at which plastic deformation may take place in 

frames, depending on the configuration, detailing, and relative strength of the beams, columns, 

and connections include: within the connection assembly itself, as is common for shear tab type 

framing connections, within the column panel zone, or within the column. The total interstory 

drift angle, as used in these Recommended Criteria is equal to the sum of the plastic drift, as 

described herein, and the elastic interstory drift. 

Plastic 

Hinges 

Deformed 

frame shape 

Undeformed 

frame 

L’ 

L 

h 

drift angle − 

q 

Figure 6-4 Inelastic Behavior of Frames with Hinges in Beam Span 

Commentary: Nonlinear deformation of frame structures is accommodated 

through the development of inelastic flexural or shear strains within discrete 

regions of the structure. At large inelastic strains these regions can develop into 

plastic hinges, which can accommodate significant concentrated rotations at 

constant (or nearly constant) load through yielding at tensile fibers and yielding 

and buckling at compressive fibers. If a sufficient number of plastic hinges 

develop in a frame, a mechanism is formed and the frame can deform laterally in 

a plastic manner. This behavior is accompanied by significant energy dissipation 

and potentially substantial damage to the highly strained elements. The 

formation of hinges in columns, as opposed to beams, is generally undesirable, as 

this may result in the formation of mechanisms with relatively few elements 

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participating, so called “story mechanisms,” and consequently little energy 

dissipation throughout the structure. 

The prequalified connection contained in the building codes prior to the 1994 

Northridge earthquake was based on the development of plastic hinges within the 

beams at the face of the column, or within the column panel zone. If the plastic 

hinge develops in the column panel zone, the resulting column deformation results 

in very large secondary stresses on the beam flange to column flange joint, a 

condition that can contribute to brittle failure. If the plastic hinge forms in the 

beam, at the face of the column, this can result in large strain demands on the 

weld metal and surrounding heat affected zones. These conditions can also lead 

to brittle failure. 

Welded steel moment-frame structures are expected to be capable of extensive 

amounts of energy dissipation through the development of plastic hinges. In 

order to achieve reliable performance of these structures, frame configurations 

should avoid a strong beam-weak column design that can lead to column hinging 

and story collapse mechanisms. Further, beam-column connections should be 

configured to force the inelastic action (plastic hinge) away from the column face, 

where its performance is less dependent on the material and workmanship of the 

welded joint. This can be done either by local reinforcement of the connection, or 

local reduction of the cross section of the beam, at a distance away from the 

connection. Plastic hinges in steel beams have finite length, typically on the order 

of half the beam depth. Therefore, the location for the plastic hinge should be 

shifted at least that distance away from the face of the column. When this is done 

through reinforcement of the connection, the flexural demands on the columns, for 

a given beam size, are increased. Care must be taken to ensure that weak column 

conditions are not inadvertently created by local strengthening of the connections. 

Many existing WSMF structures were not configured in the original design to 

produce a strong-column, weak-beam condition. In these structures, connection 

upgrades that reinforce the beam section locally at the connection, to shift the 

location of plastic hinging into the beam span, will have little effect, as plastic 

behavior of the frame will be controlled through plastic hinging of the columns. 

In such structures, upgrade should include strengthening of the columns with 

cover plating or other similar measures, or alternatively, the provision of 

supplemental lateral force resisting elements such as braced frames or shear 

walls. Upgrade recommendations are discussed in Chapter 5. 

Connection upgrades of the type described above, while believed to be 

effective in preventing brittle connection fractures, will not prevent structural 

damage from occurring. Brittle connection fractures are undesirable for several 

reasons. First, severe connection degradation can result in loss of gravity load 

carrying capacity of the framing at the connection and the potential development 

of local collapse. From a global perspective, the occurrence of many connection 

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fractures results in a substantial reduction in the lateral-force-resisting strength 

and stiffness of the structure which, in extreme cases, can result in instability and 

collapse. Connections upgraded as described in this document should experience 

many fewer brittle fractures than unmodified connections. However, the 

formation of a plastic hinge within the beam is not a completely benign event. 

Beams that have experienced significant plastic rotation at such hinges may 

exhibit large buckling and yielding deformation, as well as concurrent localized 

damage to floor slabs and other supported elements. In severe cases, this damage 

must be repaired. The cost and difficulty of such repairs could be comparable to 

the costs incurred in repairing connection fracture damage of the types 

experienced in the Northridge earthquake. The primary difference is that life 

safety protection will be significantly enhanced and most upgraded structures 

should continue to be safe for occupancy, while repairs are made. 

If the types of damage described above are unacceptable for a given building, 

then alternative upgrade systems should be considered, which will reduce the 

plastic deformation demands on the structure during a strong earthquake. 

Appropriate methods of achieving such goals include the installation of 

supplemental braced frames, shear walls, energy dissipation systems, base 

isolation systems, and similar structural systems. 

6.3.2 Required Drift Angle Capacity 

For systematic upgrade design, the required drift angle capacity of connection assemblies 

should be sufficient to withstand the total (elastic and plastic) interstory drift likely to be induced 

in the frame by earthquake ground shaking, as predicted by analysis, while providing sufficient 

confidence with regard to achievement of the desired performance, in accordance with the 

procedures of Chapter 3. Section 6.6 provides data on the drift angle capacity of several 

prequalified connection upgrade details, together with design guidelines for these connection 

upgrades and limits on the applicability of the prequalification. Section 6.7 provides 

performance data for several types of moment-resisting connections that have been prequalified 

for use in new steel moment-frame construction. Section 6.8 provides descriptive information on 

several types of proprietary connection technologies that may be considered for seismic upgrade 

applications. Section 6.9 provides recommended criteria for determining the factored drift angle 

capacity of connection upgrades that are not prequalified. 

For the purposes of Simplified Upgrade, frames shall be classified either as Ordinary Moment 

Frames (OMF) or Special Moment Frames (SMF) and connection upgrade details that are 

prequalified for the appropriate system, as indicated in Section 6.6 of these guidelines, should be 

selected. For purposes of simplified upgrades, a frame should be considered an SMF system if 

the construction documents indicate it was designed as a Special Moment Resisting Frame, a 

Ductile Moment Resisting Frame, or if the original design documents indicate that any of the 

design values indicated in the column labeled “SMF” in Table 6-4 were used in determining the 

design seismic forces for the frame in the original design. A frame should be considered an OMF 

if the design documents indicate it was designed as an OMF or if any of the design values 

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indicated in the column labeled “OMF” in Table 6-4 were used in determining the design seismic 

forces for the frame in the original design. If sufficient documentation is not available to permit 

determination of the original intended system for the structure, an SMF should be assumed. 

Table 6-4 Design Coefficients for SMF and OMF Systems 

Design Coefficient OMF SMF 

K 

(buildings designed to 1985 or earlier edition of UBC, or 1990 or earlier editions 

of BOCA or SSBC.) 

1.0 0.67 

Rw 

(buildings designed to UBC editions 1988 - 1994) 

6 12 

R 

(buildings designed to 1997 UBC, or 1993 or later editions of BOCA or SSBC.) 

4 8 

Commentary: In Systematic Upgrades, a complete analysis of the structure is 

performed, in accordance with the criteria of Chapter 3. In this analysis, an 

estimate is developed of the forces and deformations induced by response to 

earthquake ground shaking, and based on these estimated forces and 

deformations, and the estimated capacity of the frame and its individual 

components to resist these demands, a level of confidence with regard to the 

ability of the frame to provide desired performance is estimated. 

In Simplified Upgrades, performance evaluation of the structure, in 

accordance with Chapter 3, is not performed. Rather than providing a specific 

level of confidence that the structure is capable of a particular performance, 

simplified upgrades are intended only to provide the structure with the level of 

reliability implicitly presumed by the code provisions under which it was 

originally designed. Until recently, the building codes only recognized two types 

of moment-resisting steel frame systems: a system with significant intended 

inelastic response capability called either a Special Moment Frame, or in some 

codes, a Ductile Moment-Resisting Frame; and frames having only limited 

inelastic response capability, typically called an Ordinary Moment Frame. 

Table 6-4 classifies framing systems, using the terminology contained in the 

1997 NEHRP Recommended Provisions for New Buildings and 1997 AISC 

Seismic Design Specification, as either an SMF or an OMF. 

In addition to these two categories of moment-resisting frames, some steel 

moment-resisting frames are part of a dual structural system, in which the frames 

provide a secondary system of lateral-force resistance for a primary system 

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comprised of braced frames or shear walls. Upgrade of such structures, using the 

Simplified procedure is not recommended. 

6.3.3 Connection Configuration 

For Simplified Upgrade, a connection upgrade configuration should be selected that is 

compatible with the appropriate structural system. No further qualification of the design is 

necessary, other than to ensure that the connection configuration does not create any of the 

following conditions, as defined in the building code, or make an existing such condition more 

severe: 

a. Weak column - strong beam 

b. Weak story 

c. Soft story 

d. Torsional Irregularity 

For Systematic Upgrade, a connection configuration that is capable of providing sufficient 

factored drift angle capacity to provide a suitable level of confidence should be selected. Section 

6.6 presents data on a series of prequalified connection upgrade details, from which an 

appropriate detail may be selected. These connection upgrades details are prequalified for use 

within certain ranges of member sizes and frame configuration. If these connection upgrade 

details are to be employed outside the range of applicability, project specific connection 

qualification should be performed. If project-specific connection qualification is to be 

performed, a connection of any configuration may be selected and qualified for acceptability 

using the procedures of Section 6.9. 

6.3.4 Determine Plastic Hinge Locations 

Based on the data presented in these Recommended Criteria for prequalified connection 

upgrades, or data obtained from a qualification testing program for configurations that are 

qualified on a project specific basis, the location of expected plastic hinge formation, sh, as 

indicated in Figure 6-5 should be identified. The plastic hinge locations presented for 

prequalified connection upgrades are valid for beams with gravity loads representing a small 

portion of the total flexural demand and for conditions of strong column, weak beam. For frames 

in which gravity loading produces significant flexural stresses in the members, or frames that do 

not have strong-column, weak-beam configurations, locations of plastic hinge formation should 

be determined based on methods of plastic analysis. 

Commentary: The suggested location for the plastic hinge, as indicated by the 

parameter sh in the prequalification data, is valid only for frames with limited 

gravity loading present on the frame beams, or for frames in which yielding will 

actually occur in the beam, rather than in the column panel zone or the column 

itself. If significant gravity load is present, or if panel zones or columns are the 

weak links in the frame, this can shift the locations of the plastic hinges, and in 

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the extreme case, change the form of the collapse mechanism. If flexural demand 

on the girder due to gravity load is less than about 30% of the girder plastic 

capacity, this effect can safely be neglected, and the plastic hinge locations taken 

as indicated, as long as beam flexure, rather than panel zone shear, column 

flexure, or beam shear is the dominant inelastic behavior for the frame. If gravity 

demands significantly exceed this level then plastic analysis of the girder should 

be performed to determine the appropriate hinge locations. 

ıShˆ 

Beam depth − d 

L’ 

Plastic 

hinge 

Reduced beam 

section 

(if applicable) 

ıSh ˆ 

Connection 

reinforcement 

(if applicable) 

L 

Figure 6-5 Location of Plastic Hinge Formation 

6.3.5 Determine Probable Plastic Moment at Hinges 

For fully restrained connections designed to develop plastic hinging in the beam or girder, the 

probable plastic moment at the location of the plastic hinge should be determined as: 

M pr = CprRyZeFy (6-3) 

where: 

Mpr = Probable peak plastic hinge moment. 

Cpr = A factor to account for the peak connection strength, including strain 

hardening, local restraint, additional reinforcement, and other connection 

conditions. For most connection types, Cpr is given by the formula: 

F +F 

Cpr = y u (6-4) 2 Fy 

A value of 1.2 may be used for all cases, except where otherwise noted in 

the individual connection design procedures included with the 

prequalifications in later sections of these Recommended Criteria. 

Ry = A coefficient, applicable to the beam or girder material, obtained from the 

AISC Seismic Provisions 

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Ze = The effective plastic modulus of the section (or connection) at the location 

of the plastic hinge. 

Fy = the specified minimum yield stress of the material of the yielding element. 

Fu = the specified minimum tensile stress of the material of the yielding 

element. 

For connections that do not develop plastic hinges in the beam, the hinge strength should be 

calculated, or determined from tests, for the pertinent yield mechanism, considering the variation 

in material properties of the yielding elements. For prequalified connection upgrades and 

connections, calculation methods to determine the yield strengths of the various active 

mechanisms are given in the design procedure accompanying the individual prequalification. 

Commentary: The AISC Seismic Provisions use the formulation 1.1RyMp+Mv for 

calculation of the quantity SM*pb, which is used in calculations for column 

strength (strong-column, weak-beam), and for required shear strength of panel 

zones. As described in FEMA-355D, research has shown that, for most 

connection types, the peak moment developed is somewhat higher than the 1.1 

factor would indicate. Therefore, for these guidelines, the factor Cpr , calculated 

as shown, is used for individual connections, with a default value of 1.2 

applicable to most cases. 

6.3.6 Determine Shear at the Plastic Hinge 

The shear at the plastic hinge should be determined by statics, considering gravity loads 

acting on the beam. A free body diagram of that portion of the beam between plastic hinges is a 

useful tool for obtaining the shear at each plastic hinge. Figure 6-6 provides an example of such 

a calculation. For the purposes of such calculations, gravity load should be based on the load 

combinations indicated in Section 6.5.1. 

6.3.7 Determine Strength Demands at Each Critical Section 

In order to complete the design of the connection upgrade, including, for example, sizing the 

various plates, bolts, and joining welds, which make up the connection, it is necessary to 

determine the shear and flexural strength demands at each critical section. These demands may 

be calculated by taking a free body of that portion of the connection assembly located between 

the critical section and the plastic hinge. Figure 6-7 demonstrates this procedure for two critical 

sections for the beam shown in Figure 6-6. 

Commentary: Each unique connection configuration may have different critical 

sections. The vertical plane that passes through the joint between the beam 

flanges and column (if such joining occurs) will typically define at least one such 

critical section, used for designing the joint of the beam flanges to the column, as 

well as evaluating shear demands on the column panel zone. A second critical 

section occurs at the center line of the column. Moments calculated at this point 

are used to check strong-column, weak-beam conditions. Other critical sections 

should be selected as appropriate. 

6-16
Mpr

taking the sum of moments about ıAˆ = 0

VA

' 

' 2 ' 2 2 

L 

M M P L W L

Vp pr + pr + + = 

M pr

taking the sum of moments about ıAˆ  0

V A

Plastic 

Vp

Mpr

Sh−dc /2 

Mf

Sh 

dc 

M

c

Critical Section at Column Face Critical Section at Column Centerline 

= + M f pr VS ph = M c M pr 

Vp

Mpr

Plastic 

Vp

Mpr

Sh−dc /2 

Mf

Plastic 

Vp

Mf Mpr

Sh 

dc 

M

c

Critical Section at Column Face Critical Section at Column Centerline 

= + M f pr VS ph = M c M pr 

Vp

Mpr

dc

M

c

Vp

Mpr

Vp

M - +V p

Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings Chapter 6: Connection Qualification 

L

L sh ’ 

Plastic Plastic

hingeP

L/2

P

Mpr

VpL’

ıAˆ

w

L

L sh ’ 

Plastic Plastic

hingeP

L/2

P

Mpr

VpL’

ıAˆ

w

V A 

M pr 

taking the sum of moments about ıAˆ = 0 

2 '

2

pr + pr + ' + 2

Vp 

M M P L W L = L' 

Figure 6-6 Sample Calculation of Shear at Plastic Hinge 

Plastic Plastic 

hingehinge 

M M f pr M M pr c

Vp Vp dc 

Sh−dc /2 Sh 

2 M f = M pr +VS ph - d cM c = M pr +Vp

Critical Section at Column Face Critical Section at Column Centerline 

Figure 6-7 Calculation of Demands at Critical Sections 

6.3.8 Yield Moment 

The design procedures for some prequalified connections contained in these Recommended 

Criteria require that the moment at the face of the column at onset of plastic hinge formation, 

Myf, be determined. Myf may be determined from the following equation: 

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M yf = CyM f (6-5) 

where: 

1 C = (6-6) y Z

C be 

pr Sb 

Cpr = the peak connection strength coefficient defined in Section 6.3.5 

Sb = the elastic section modulus of the beam at the zone of plastic hinging 

Zbe = the effective plastic section modulus of the beam at the zone of plastic hinging. 

6.4 General Requirements 

This section provides criteria for connection upgrade design conditions that are considered to 

be general, that is, those conditions which, when they occur in a connection upgrade, are 

considered to perform in a similar way, or at least to have the same requirements for successful 

performance, irrespective of the connection type being used. The designer should employ these 

criteria in the design of all connection types, except when specific testing has been performed 

that qualifies the connection for use with different conditions, or unless otherwise specifically 

indicated in these Recommended Criteria. 

6.4.1 Framing 

6.4.1.1 Beam and Column Strength Ratio 

For multistory SMF systems, frames should be configured with a strong-column, weak-beam 

configuration, to avoid the formation of single-story mechanisms. As a minimum, Equation 9-3 

of AISC Seismic Provisions should be satisfied. In the application of Equation 9-3, the quantity 

Mc as defined in Section 6.3.7 of these Recommended Criteria should be substituted for the 

* quantity M pb . 

Commentary: When subjected to strong ground shaking, multi-story structures 

with columns that are weaker in flexure than the attached beams can form single 

story mechanisms, in which plastic hinges form at the base and top of all columns 

in a story. Once such a mechanism forms in a structure, nearly all of the 

earthquake induced lateral displacement will occur within the yielded story, 

which can lead to very large local drifts and the onset of P-D instability and 

collapse. 

Building codes permitted frames to be designed with weak-column, strongbeam configurations until 1988. Therefore, many existing steel moment-frame 

buildings have such configuration. Further, some types of connection upgrades, 

through local strengthening of the beam ends, have the potential to create weak6-18
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Steel Moment-Frame Buildings Chapter 6: Connection Qualification 

column, strong-beam systems in frames that originally did not have such 

configuration. Although weak-column, strong-beam designs are not desirable, 

AISC Seismic does permit their use under certain conditions, even for SMF 

systems. Before utilizing weak-column, strong-beam configurations, designers 

should be aware that the prequalified connections for SMF systems contained in 

these Recommended Criteria are based on tests using strong columns. 

Nonlinear analyses of representative frames have clearly shown that the use 

of the provisions described above will not completely prevent plastic hinging of 

columns. This is because the point of inflection in the column may move away 

from the assumed location at the column mid-height once inelastic beam hinging 

occurs, and because of global bending induced by the deflected shape of the 

building, of which the column is a part. 

Except for the case when a column hinge mechanism forms, column hinging is 

not a big problem, provided that the columns are designed as compact sections, 

are properly braced and axial loads are not too high. It is well understood that a 

column hinge will form at the base of columns that are continuous into a 

basement, or that are rigidly attached to a stiff and strong foundation. 

6.4.1.2 Beam Flange Stability 

Beam flange slenderness ratios bf /2tf (b/t) should be limited to a maximum value of 52/�Fy, 

as required by AISC Seismic Provisions. For moment frame beams with Reduced Beam Section 

(RBS) connections, it is recommended that the bf /2tf be determined based on the flange width bf 

measured at the ends of the center 2/3 of the reduced section of the beam unless gravity loads are 

large enough to shift the hinge point significantly from the center point of the reduced section. 

Commentary: The AISC Seismic Provisions require that beam flange slenderness 

ratios bf /2tf (b/t) be limited to a maximum of 52/�Fy. This specific value is 

intended to allow some plastic rotation of the beam to occur before the onset of 

local buckling of the flanges, a highly undesirable phenomenon. Widespread 

buckling of beam flanges in a moment resisting frame can result in development 

of frame strength degradation increasing both story drifts and the severity of P-D 

effects and therefore should be avoided. Local flange buckling results in very 

large local straining of the flanges and the early on-set of low-cycle fatigue 

induced tearing of the beam flanges, which ultimately limits the ability of the 

assembly to withstand cyclic inelastic rotation demands. Further, severely 

buckled beam flanges can be even more difficult to repair than fractured beam 

connections. 

Notwithstanding the above, under large plastic rotation demands, buckling of 

beam flanges will inevitably occur. The value of the b/t of the beam involved in a 

specific connection can have a major effect on how the beam column assembly 

performs. Beams and girders used in moment frames should comply with the 

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limits specified by the AISC Seismic Provisions, except as specifically modified by 

individual connection prequalifications or qualification tests. 

6.4.1.3 Beam Web Stability 

Moment-frame beams should be selected that have web height-to-thickness ratios, hc/tw of not 

greater than 418/�Fy. 

Commentary: The AISC Seismic Provisions permits use of beams with web hc/tw 

up to as high as 520/�Fy, for beams without axial load. Most of the testing under 

this project has been conducted on beams such as W30x99 and W36x150, both of 

which barely conform to hc/tw< 418/�Fy. Since many of the specimens exhibited 

significant web buckling in the area of plastic hinges, it is not considered prudent 

to utilize beams with relatively thinner webs in moment frames. Although 

stiffening of the webs could be done to limit web buckling, it is possible that 

stiffeners could be detrimental to connection performance. Since connections with 

web stiffeners were not tested, such connections have not been prequalified. See 

FEMA-355D, State of the Art Report on Connection Performance, for further 

discussion of web buckling of moment-frame beams. 

6.4.1.4 Beam Span and Depth Effects 

The performance of moment-resisting beam-column connections is strongly related both to 

beam depth and beam span-to-depth ratio. Data accompanying each of the prequalified 

connection upgrades presented in Section 6.6 includes specification of maximum beam depths 

and minimum beam span-to-depth ratio. Connection upgrade details presented in Section 6.6 

should not be used for cases where beam depth exceeds the indicated limit unless project-specific 

qualification, in accordance with Section 6.9 is performed. For Simplified Upgrade, connection 

upgrade details should not be used in cases where the beam span-to-depth ratio is less than the 

indicated amount unless project-specific qualification, in accordance with Section 6.9, is 

performed. For Systematic Upgrade, connection upgrade details may be used on beams with 

spans that have smaller span-to-depth ratio than the limiting value indicated in the 

prequalification provided that the acceptance criteria used in performance evaluation for 

interstory drift capacity q as limited by local connection behavior is modified as indicated by the 

equation: 

8d æ L L - ¢ö

÷ q ¢= ç

ç1+ ÷q (6-7) L ÷ è L ø 

where: 

q' = the median interstory drift angle capacity for connection behavior for beams with 

small span-to-depth ratio 

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Steel Moment-Frame Buildings Chapter 6: Connection Qualification 

q = the median interstory drift angle capacity listed in the prequalification for 

connection behavior for beams meeting the span to depth limitations of the 

prequalification 

L = the span of the beam, center-line-to-center-line of columns, inches 

L' = the effective span of the beam between plastic hinge locations, inches 

d = the beam depth in inches 

Where the effective span L' of the beam between points of plastic hinging, is such that shear 

yielding of the beam will occur, rather than flexural yielding, the web of the beam should be 

stiffened between the points of plastic hinging, and braced as required by the 1997 AISC Seismic 

Provisions for long links in eccentric braced frames. 

Commentary: Both beam depth and beam span-to-depth ratio are significant in 

the inelastic behavior of beam-column connections. At a given induced curvature, 

deep beams will undergo greater straining than shallower beams. Similarly, 

beams with shorter span-to-depth ratio will have a sharper moment gradient 

across the beam span, resulting in reduced length of the beam participating in 

plastic hinging and increased strains under inelastic rotation demands. Most of 

the beam-column assemblies tested under this project used configurations 

approximating beam spans of about 25 feet and beam depths varying from W30 to 

W36 so that beam span-to-depth ratios were typically in the range of 8 to 10. 

Equation 6-7 approximately accounts for these effects. Additional information 

may be found in FEMA-355D, State of the Art Report on Connection 

Performance. 

6.4.1.5 Beam Flange Thickness Effects 

The connection upgrade prequalifications contained in these Recommended Criteria are 

limited in application to specific beam flange thicknesses. These limitations are noted in the 

tabulated data for each connection. For frames designed using project-specific connection 

qualifications, connection tests used in the connection qualification program should employ 

beam flanges of similar or greater thickness than those used in the frame. 

Commentary: In addition to controlling the stability of the flange under 

compressive loading, as described above, beam flange thickness also affects the 

size of welds in welded connections. Although it is not a given that larger welds 

will be less reliable than smaller welds, greater control may be necessary to 

ensure their performance, and quality control may be more difficult. Additionally, 

residual stresses are likely to be higher in thicker material with thicker welds. 

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6.4.1.6 Lateral Bracing at Beam Flanges at Plastic Hinges 

Plastic hinge locations that are remote from the column face in beams that do not support a 

slab should be provided with supplemental bracing, as required by the 1997 AISC Seismic 

Provisions. Where the beam supports a slab and is in direct contact with the slab along its span 

length, supplemental bracing need not be provided. 

Commentary: The 1997 AISC Seismic Provisions require that beam flanges be 

braced at plastic hinge locations. Because plastic hinges have been moved away 

from the column face for some of the connection upgrade types in this section, a 

strict interpretation of the provisions would lead to a requirement that flanges at 

such hinges be laterally braced. Limited testing conducted as part of this project 

(FEMA-355D) suggests that, as long as the hinging beam is connected to a 

concrete slab, excessive strength deterioration due to lateral buckling will not 

occur within the ranges of drift angle normally considered important. Therefore, 

these Recommended Criteria do not require supplemental bracing of plastic hinge 

locations adjacent to column connections of beams supporting slabs. 

For those cases where supplemental bracing of beam flanges near plastic 

hinges is appropriate, great care must be taken in detailing and installation of 

such bracing to ensure that attachments are not made directly within the area of 

anticipated plastic behavior. This is because of the inherent risk of reducing 

plastic deformation capacity for the beam by introducing stress concentrations or 

metallurgical notches into the region of the beam that must undergo plastic 

straining. See FEMA-355D, State of the Art Report on Connection Performance, 

for further discussion of flange bracing. 

6.4.1.7 Welded Shear Studs 

Welded shear studs, or other attachments for composite action with slabs or for diaphragm 

shear transfer, should not be installed within the hinging area of moment-frame beams. The 

hinging area is defined as the distance from the column flange face to one half the beam depth 

beyond the theoretical hinge point. Standard arc-spot weld attachments may be made in the 

hinging area, but shot-in, or screwed attachments should not be permitted. 

Commentary: It has been shown in some tests that welded shear studs and the 

rapid increase of section caused by composite action can lead to beam flange 

fractures when they occur in the area of the beam flange that is undergoing large 

cyclic strains. It is not certain whether the welding of the studs, the composite 

action, or a combination of the two is the cause, but, based on the limited 

evidence, it is judged to be prudent to permit no studs in the hinging area. It is 

also prudent to permit no attachments that involve penetration of the flanges in 

the hinging region. 

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6.4.2 Welded Joints 

6.4.2.1 Through-Thickness Strength 

The through-thickness strength demands on existing column material should be limited to the 

values given in Table 6-5. Through-thickness demands should be calculated as the applied 

flange force, divided by the projected area of the welded joint on the column flange, using the 

procedures of Section 6.3.7 to calculate the applied force at this critical section. 

Table 6-5 Column Flange Through-Thickness Strength 

Column Flange Material Specification Ft-t 

Hot rolled wide flange columns conforming to A36, ASTM A572 Grade 50, or 

ASTM A992, or ASTM A913 rolled later than 1994 and having sulfur content 

not in excess of 0.05% by weight. 

No limit 

All other material 0.8Fu 

Commentary: Early investigations of connection fractures in the 1994 Northridge 

earthquake identified a number of fractures (types C3 and C5 Section 2.3.2) that 

appeared to be the result of inadequate through-thickness strength of the column 

flange material. As a result of this, in the period immediately following the 

Northridge earthquake, a number of recommendations were promulgated that 

suggested limiting the value of through thickness stress demand on column 

flanges to a value of 40 ksi, applied to the projected area of the beam flange 

attachment. This value was selected to ensure that through-thickness yielding did 

not initiate in the column flanges of FR connections and often controlled the 

overall design of a connection subassembly. 

It is important to prevent the inelastic behavior of connections from being 

controlled by through-thickness yielding of column flanges. This is because it 

would be necessary to develop very large local ductilities in the column flange 

material in order to accommodate even modest plastic rotation demands on the 

assembly. However, the actual cause for the type C3 fractures, that were initially 

identified as through-thickness failures of the column flange are now believed to 

be unrelated to this material property. Rather, it appears that C3 damage 

occurred when fractures initiated in defects present in the complete joint 

penetration (CJP) weld root, not in the flange material (FEMA-355E). These 

defects sometimes initiated a crack, that under certain conditions, propagated 

into the column flange, giving the appearance of a through-thickness failure. 

Detailed fracture mechanics investigations conducted under this project confirm 

that the C3 damage initially identified as through-thickness failures are likely to 

have occurred as a result of certain combinations of material strength and notch 

toughness, conditions of stress in the connection, and the presence of critical 

flaws in the welded joint. 

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As part of the research conducted in support of the development of these 

Recommended Criteria, extensive through-thickness testing of modern steels, 

meeting the ASTM A572, Gr. 50 and ASTM A913, Gr. 65 specifications has been 

conducted to determine the susceptibility of modern column materials to throughthickness failures (FEMA 355A, State of the Art Report on Base Metals and 

Fracture). This combined analytical and laboratory research clearly showed that 

due to the restraint inherent in welded beam flange to column flange joints, the 

through thickness yield and ultimate strengths of the column material is 

significantly elevated in the region of the connection. Further, for the modern 

materials tested, these strengths significantly exceed those that can be delivered 

to the column by beam material conforming to these same specifications. For this 

reason, no limits are suggested for the through-thickness strength of modern steel 

materials with controlled sulfur contents, as required by the FEMA-353 

Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications. 

Notwithstanding the above, it is known that in the past, lamellar tearing of 

thick column flanges occasionally occurred during the fabrication and erection 

process. This lamellar tearing was a result of high through thickness strains 

induced by welding on material that had excessive sulfur inclusions. These sulfur 

inclusions, which were flattened and elongated during the shape rolling process 

could form planes of weakness within the shape that were susceptible to this 

tearing. It is known that steel with relatively high sulfur content is more 

susceptible to this behavior than shapes with lower sulfur contents. Also, it is 

known that shapes that undergo a significant amount of working during the 

rolling process are more susceptible as well, as the rolling process tends to flatten 

the sulfide inclusions and align them in the rolling direction. Modern steel 

production often uses a continuous casting process in which the steel is cast in a 

shape that is near that of the final product, resulting in the sulfur being uniformly 

distributed throughout the shape and therefore less susceptibility to lamellar 

tearing. 

Table 6-5 recommends a limit of 0.8Fu for through-thickness stress on older 

steels, that may be susceptible to through-thickness tearing, based on a statistical 

survey of the relationship of through-thickness strength to longitudinal strength 

for structural steels (Barsom, 1996). 

6.4.2.2 Base Material Toughness 

Material in rolled shapes with flanges 1-1/2 inches or thicker, and sections made from plates 

that are 2 inches or thicker, should be required to have minimum Charpy V-notch toughness of 

20 ft-lbs, at 70 degrees F. Refer to FEMA-353, Recommended Specifications and Quality 

Assurance Guidelines for Steel Moment-Frame Construction for Seismic Applications. 

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Commentary: The 1997 AISC Seismic Provisions specified minimum notch 

toughness for rolled shapes with flanges 1-1/2 inches thick or thicker, and 

sections made from plates 1-1/2 inches thick or thicker, be checked for notch 

toughness. These Recommended Criteria relax the requirement for toughness of 

plate material to apply to plates 2 inches or thicker as this was the original intent 

of the AISC specification, and it is believed that the AISC document will be 

revised to this requirement. 

Research has not clearly demonstrated the need for a specific value of base 

metal toughness. However, it is judged that base metal notch toughness is 

important to prevention of brittle fracture of the base metal in the highly stressed 

areas of the connection. A number of connection assemblies that have been tested 

have demonstrated base metal fractures at weld access holes and at other 

discontinuities such as at the ends of cover plates. In at least some of these tests, 

the fractures initiated in zones of low notch toughness. Tests have not been 

conducted to determine if higher base metal notch toughness would have reduced 

the incidence of such fractures. 

The Charpy V-Notch (CVN) value of 20 ft.-lbs. at 70 degrees F, recommended 

here, was chosen because it is usually achieved by modern steels, and because 

steels meeting this criterion have been used in connections which have performed 

successfully. Current studies (FEMA 355A, State of the Art Report on Base 

Metals and Fracture) have indicated that rolled shapes produced from modern 

steels meet this requirement almost routinely even in the thicker shapes currently 

requiring testing. It has been suggested that the requirement for this testing could 

be eliminated and replaced by a certification program administered by the mills. 

However, such a program is not currently in existence. Until such time as such a 

certification program is in place, or a statistically meaningful sampling from all 

major mills has been evaluated, it is recommended that the AISC requirement for 

testing be continued. According to the Commentary to the 1997 AISC Seismic 

Provisions, thinner sections are judged not to require testing because they “are 

generally subjected to enough cross-sectional reduction during the rolling 

process that the resulting notch toughness will exceed that required.” In other 

words, the notch toughness is required, but testing to verify it on a project basis is 

not judged to be necessary as it is routinely achieved. 

No specific notch toughness requirements are specified for existing materials 

in steel moment frames. This is because testing of the notch toughness of these 

materials is costly and difficult and also because there is no practical way to 

improve the notch toughness of an existing material, other than to replace it. The 

importance of base material notch toughness with regard to steel moment-frame 

behavior is not clear, however. High material notch toughness is beneficial in 

preventing the propagation of minor fractures and flaws into unstable brittle 

fractures, when such defects are present. However, base metals typically are free 

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of such defects and therefore, less susceptible to the initiation of the brittle 

fractures that material notch toughness is effective in preventing. 

6.4.2.3 k-Area Properties 

The k-area of rolled wide-flange shapes, which may be considered to extend from the midpoint of the radius of the fillet from the flange into the web, approximately 1 to 1-1/2 inches 

beyond the point of tangency between the fillet and web, as defined in Figure C-6.1 of the AISC 

Seismic Provisions, is likely to have low toughness and may therefore be prone to cracking 

caused by welding operations. Designers should detail welds of continuity plates and web 

doubler plates in columns in such a way as to avoid welding directly in the k-area. Refer to 

Section 6.4.3 for more information. 

Fabricators should exercise special care when making welds in, or near to, the k-area. Where 

welding in the k-area of columns cannot be avoided, special nondestructive testing is 

recommended. Refer to FEMA-353, Recommended Specifications and Quality Assurance 

Guidelines for Steel Moment-Frame Construction for Seismic Applications. 

Commentary: Recent studies, instigated in response to fabrication problems, have 

shown that, for rotary-straightened W-shapes, an area of low material toughness 

can occur in the region of the web immediately adjacent to the flange. In some 

instances, cracking has occurred in these areas during welding. The Commentary 

to the AISC Seismic Provisions provides a figure (Fig. C-6.1) that defines the karea. 

The low toughness of the k-area seems to be associated only with rotarystraightened sections. Which sections are rotary straightened varies among the 

mills. One major domestic supplier rotary-straightens all shapes weighing less 

than 150 pounds per linear foot. Larger sections are often straightened by other 

means that do not result in as much loss of toughness in the k-area. Because 

rolling practice is frequently changed, it is prudent to assume that all rolled 

sections are rotary-straightened. 

6.4.2.4 Weld Filler Metal Matching and Overmatching 

The use of weld filler metals and welding procedures that will produce welds with matching 

or slightly overmatching tensile strength relative to the connected steel is recommended. 

Welding consumables specified for Complete Joint Penetration (CJP) groove welds of beam 

flanges and flange reinforcements should have yield and ultimate strengths at least slightly higher 

than the expected values of yield and ultimate strength of the beam or girder flanges being 

welded. Significant overmatching of the weld metal should not be required unless overmatching 

is specified in the connection prequalification or is used in the prototypes tested for projectspecific qualification of the connection being used. Flux Cored Arc Welding and Shielded Metal 

Arc Welding electrodes commonly used in structural construction and conforming to the E70 

specifications provide adequate overmatching properties for structural steels conforming to 

ASTM A36, A572, Grades 42 and 50, A913, Grade 50 and A992. Welded splices of columns of 

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A913-Grade 65 steel should be made with electrodes capable of depositing weld metal with a 

minimum ultimate tensile strength of 80 ksi. 

Commentary: Undermatched weld metals, that is, weld metals with lower strength 

than the connected base metals, are beneficial in some applications in that they 

tend to limit the residual stress state in the completed joint. However, in 

applications where yield level stresses are anticipated, it is desirable to minimize 

the amount of plasticity in the welded joint. This can be achieved by employing 

balanced, or slightly overmatched weld filler metals. The majority of the 

successful connection tests have used weld metals with yield and tensile strengths 

in the range of 58 and 70 ksi respectively, which provide matching to moderate 

overmatching with beams of Grade 50 steel. For additional information refer to 

FEMA-355B, State of the Art Report on Welding and Inspection. 

6.4.2.5 Weld Metal Toughness 

For structures in which the steel frame is normally enclosed and maintained at a temperature 

of 50oF or higher, critical welded joints in seismic force resisting systems, including complete 

joint penetration (CJP) groove welds of beam flanges to column flanges, CJP welds of shear tabs 

and beam webs to column flanges, column splices, and similar joints, should be made with weld 

filler metal providing CVN toughness of 20ft-lbs at -20� F and 40ft-lbs at 70� F and meeting the 

Supplemental Toughness Requirements for Welding Materials in FEMA-353 – Recommended 

Specifications and Quality Assurance Guidelines for Steel Moment-Frame Construction for 

Seismic Applications. For structures with lower service temperatures than 50o

F, qualification 

temperatures should be reduced accordingly. 

Commentary: Principles of fracture mechanics demonstrate the importance of 

notch toughness to resist fracture propagation from flaws, cracks, and backing 

bars or other stress concentrations, which may be preexisting or inherent, or 

which may be caused by applied or residual stresses. The 1997 AISC Seismic 

Provisions requires the use of welding consumables with a rated Charpy V-Notch 

(CVN) toughness of 20 ft.-lbs. at -20� F, for CJP groove welds used in the Seismic 

Force Resisting System. Seismic Provisions for Structural Steel Buildings (1997) 

Supplement No. 1, February 15, 1999, (AISC, 1999) changes this requirement to 

include “all welds used in primary members and connections in the Seismic Force 

Resisting System”. The rating of the weld filler metal is as determined by the 

American Welding Society classification or manufacturer certification. 

Studies conducted under this project have indicated that not all weld 

consumables that are rated for 20 ft-lbs of toughness at –20o

F will provide 

adequate toughness at anticipated service temperatures. The supplemental 

toughness requirements contained in FEMA-353 are recommended to ensure that 

weld metal of adequate toughness is obtained in critical joints. Most of the beamcolumn connection tests conducted under this project were made with weld filler 

metal conforming to either the E70T6 or E70TGK2 designations. These filler 

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metals generally conform to the recommended toughness requirements. Other 

weld filler metals may also comply. 

6.4.2.6 Weld Backing, Weld Tabs, and other Welding Details 

Weld backing and runoff tabs should be removed from complete joint penetration flange 

welds, unless otherwise noted in the connection prequalification or demonstrated as not required 

by project-specific qualification testing. Refer to FEMA-353, Recommended Specifications and 

Quality Assurance Guidelines for Steel Moment-Frame Construction for Seismic Applications, 

for special requirements for weld backing, weld tabs, and other welding details for moment 

frame joints. It is not recommended that backing and runoff tabs be removed from existing 

connections in buildings, unless other upgrades or modifications of the affected connections are 

being made, in which case such removal is recommended. 

The following general procedures may be considered for backing removal. Steel backing 

may be removed either by grinding or by the use of air arc or oxy-fuel cutting. The zone just 

beyond the theoretical 90-degree intersection of the beam-to-column flange should be removed 

either by air arc or oxy-fuel cutting followed by a thin grinding disk, or by a grinding disk alone. 

This shallow gouged depth of weld and base metal should then be tested by magnetic particle 

testing (MT) to determine if any linear indications remain. If the area is free of indications the 

area may then be re-welded. The preheat should be maintained and monitored throughout the 

process. If no further modification is to be made or if the modification will not be affected by a 

reinforcing fillet weld, the reinforcing fillet may be welded while the connection remains at or 

above the minimum preheat temperature and below the maximum interpass temperature. 

Commentary: It was originally hypothesized, following the 1994 Northridge 

earthquake that weld backing created an effective crack equal to the thickness of 

the backing and that this phenomena was responsible for many of the fractures 

that had occurred. Finite-element analyses of welded joints (Chi, et al., 1997) 

have shown that although the backing does create some notch effect, a far more 

significant factor is the fact that when backing is left in place, it obscures effective 

detection of significant flaws that may exist at the weld root. These flaws 

represent a significantly more severe notch condition than does the backing itself. 

In new construction, as stated in FEMA-353, Recommended Specifications 

and Quality Assurance Guidelines for Steel Moment-Frame Construction for 

Seismic Applications, or in modification of existing joints conducted as part of an 

upgrade project, it is recommended that backing be removed from beam bottom 

flange joints, to allow identification and correction of weld root flaws. This is not 

recommended for top flange joints because the stress condition at the top flange is 

less critical and less likely to result in initiation of fracture, even if some weld 

root flaws are present. Also, as a result of position, it is far less likely that 

significant flaws will be incorporated in top flange joints. 

Weld tabs represent another source of discontinuity at the critical weld 

location. Additionally, the weld within the weld tab length is likely to be of lower 

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quality and more prone to flaws than the body of the weld. Flaws in the weld tab 

area can create stress concentrations and crack starters and for this reason their 

removal is recommended. It is important that the process of removal of the runoff 

tabs not be, of itself, a cause of further stress concentrations, and therefore, 

FEMA-353 recommends that the workmanship result in smooth surfaces, free of 

defects. 

Removal of existing backing and weld tabs as a sole means of building 

upgrade is not recommended. Laboratory testing demonstrates that existing 

unreinforced welded type FR connections made with low notch toughness weld 

metal are incapable of ductile performance, even with the removal of these stress 

rising features. However, they should be removed as part of any program of more 

substantial upgrades of connections. 

6.4.2.7 Reinforcing Fillet Welds and Weld Overlays 

When weld backing is removed, the weld should be reinforced with a fillet weld. The size of 

the weld should be sufficient to cover the root of the existing Complete Joint Penetration weld, 

and not less than 1⁄4 -in. The profile of the fillet should be as described in Section 5.4 of AWS 

D1.1 with a transition free from undercut, except as permitted by AWS D1.1. 

One method for improving the performance of existing unreinforced connections with low 

notch toughness weld metal is to reinforce the existing welded joints with weld overlays. This 

method, which is described in FEMA-352 Recommended Postearthquake Evaluation and Repair 

Criteria for Welded Steel Moment-Frame Buildings, is not prequalified for any specific 

performance capability, though it is known to be capable of some significant performance 

improvement. 

Commentary: Limited testing on the use of built-up welds (overlay welds) as a 

means of repairing and reinforcing welded connections of smaller-sized beams in 

existing buildings has been performed. This upgrade technique has not been 

prequalified with regard to performance capability as insufficient laboratory test 

data are available at this time to qualify its use and provide the necessary 

statistical data on its performance. 

6.4.2.8 Weld Access Hole Size, Shape, Workmanship 

New welded moment-resisting connections should utilize weld-access hole configurations as 

shown in Figure 6-8, except as otherwise noted in specific details in these Recommended 

Criteria. Criteria for cutting and finishing of weld access holes are provided in FEMA-353, 

Recommended Specifications and Quality Assurance Guidelines for Steel Moment-Frame 

Construction for Seismic Applications. 

Commentary: The size, shape, and workmanship of weld-access holes can affect 

connection strength in several different ways. If the hole is not large enough, this 

restricts welder access to the joint and increases the probability of low quality 

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joints. Depending on the size and shape of the weld access hole plastic strain 

demands in the welded joint and in the beam flange at the toe of the weld access 

hole can be significantly affected. Laboratory tests of unreinforced connections 

fabricated with tough weld filler metals have indicated that these connections 

frequently fail as a result of low cycle fatigue of the beam flange material at the 

toe of the weld access hole, as a result of the strain concentrations introduced by 

this feature. The configuration shown in Figure 6-8 was developed as part of the 

program of research conducted under this project and appears to provide a good 

balance between adequate welder access and minimization of stress and strain 

concentration. For further discussion of weld access holes, see FEMA-355D, 

State of the Art Report on Connection Performance. 

6.4.2.9 Welding Quality Control and Quality Assurance 

FEMA-353, Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications, contains recommendations for quality 

control and quality assurance for steel moment frames and connections intended for seismic 

applications. Recommended inspections are divided into two categories: Process and Visual 

Inspection, and Nondestructive Testing. For each category, different levels of inspection are 

specified depending on the anticipated severity of loading, or demand (Seismic Weld Demand 

Category) and the consequences of welded joint failure (Seismic Weld Consequence Category). 

All welded joints in the Seismic Force Resisting System should be categorized according to the 

applicable Consequence and Demand Categories, using the following form: “QC/QA Category 

BH/T”, where the first letter (in this case B) indicates the Demand Category, the second letter (in 

this case H) indicates the Consequence Category and the third letter, either T or L indicates that 

primary loading is either transverse or longitudinal, respectively. The various categories are 

described in detail in the referenced document. For the prequalified connection upgrades 

described in these Recommended Criteria, the appropriate categories have been preselected and 

are designated in information accompanying the prequalification. 

Commentary: FEMA-353 describes the Demand(A,B,C) and Consequence 

(H,M,L) Categories and indicates the appropriate levels of Visual and 

nondestructive testing (NDT) inspection for each combination of demand and 

consequence. The degree of inspection recommended is highest for the 

combination of high demand (Category A) with high consequence (Category H) 

and, conversely, less inspection is required for low demand (Category C) with low 

consequence (Category L). Intermediate degrees of inspection apply for 

intermediate categories. 

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6-31 

Tolerances shall not 

accumulate to the 

extent that the angle 

of the access hole cut 

to the flange surface 

exceeds 25�. 
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Notes: 

1. Bevel for groove weld selected. 

2. Larger of tbf or 1⁄2 inch (plus 1⁄2 tbf , or minus 1⁄4 tbf). 

3. 3⁄4 tbf to tbf . 3⁄4” min (– 1⁄4 inch). 

4. 3/8” min. radius (plus not limited, or minus 0) 

5. 3 tbf (– 1⁄2 inch). 

6. See FEMA-353, Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications, for fabrication details including 

cutting methods and smoothness requirements. 

Figure 6-8 Recommended Weld Access Hole Detail 

6.4.3 Other Design Issues for Welded Connections 

6.4.3.1 Continuity Plates 

Unless project-specific connection qualification testing is performed to demonstrate that 

beam flange continuity plates are not required, moment-resisting connections should be provided 

with beam flange continuity plates across the column web when the thickness of the column 

flange is less than the value given either by Equation 6-8 or 6-9: 

tcf < 0.4� (1.8bf tf Fyb / Fyc) (6-8) 

tcf < bf / 6 (6-9) 

where: 

tcf = minimum required thickness of column flange when no continuity plates 

are provided, inches 

bf = beam flange width, inches 

tf = beam flange thickness, inches 

Fyb = minimum specified yield stress of the beam flange, ksi 

Fyc = minimum specified yield stress of the column flange, ksi 

Where continuity plates are required, the thickness of the plates should be determined 

according to the following: 

• For one-sided (exterior) connections, continuity plate thickness should be at least one-half of 

the thickness of the beam flanges. 

• For two-sided (interior) connections, the continuity plates should be equal in thickness to the 

thicker of the two beam flanges entering the connection on either side of the column. 

• The plates should also conform to Section K1.9 of AISC-LRFD Specifications. 

Continuity plates should be welded to column flanges using complete joint penetration (CJP) 

welds as shown in Figure 6-9. Continuity plates should be welded to the web, as required, to 

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transmit the shear forces corresponding to development of the axial strength — of the CJP weld 

at one end of the connection, for one-sided connections, and that at both ends, for two-sided 

connections. 

Commentary: Following the 1994 Northridge earthquake, some engineers 

postulated that the lack of continuity plates was a significant contributing factor 

to the failure of some connections. This was partially confirmed by initial tests 

conducted in 1994 in which several specimens without continuity plates failed 

while some connections with these plates successfully developed significant 

ductility. Based on this, FEMA-267 recommended that all connections be 

provided with continuity plates. The AISC Seismic Provisions (AISC, 1997), 

which was published after FEMA-267, relaxed this criteria and states that 

continuity plates should be provided to match those in connections tested to 

obtain qualification. 

Research conducted by this project tends to confirm that where the flange 

thickness of columns is sufficiently thick, continuity plates may not be necessary. 

Equation 6-8 was the formula used by AISC to evaluate column flange continuity 

plate requirements prior to the 1994 Northridge earthquake. It appears that this 

formula is adequate to control excessive column flange prying provided that the 

beam flanges are not too wide. Studies reported in FEMA-355D suggest that the 

ratio of beam flange width to column flange thickness is also important. Tests 

with a ratio of 5.3 (W36x150 beam with W14x311 column) showed little 

difference in performance with or without continuity plates, while tests with a 

ratio of 6.8 (W36x150 beam with W27x258 column) showed some difference of 

performance. The factor of 6 in Equation 6-9 was selected by judgment based on 

these tests. 

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Notes 

1. Web doubler plate where required by Section 6.4.3.2. See the AISC Seismic Provisions Section 9.3c, 

Commentary C9.3, and Figures C-9.2 and C-9.3 for options and connection requirements. QC/QA 

Category BL/L requirement for all welds. 

2. Continuity plate as required per 6.4.3.1. 

3. Required total weld strength = ( ) pl pl Lnet Fy 0 .6t . QC/QA Category BL/L. 

4. CJP typical. QC/QA Category BM/T. 

5. AISC minimum continuous fillet weld under backing. 

6. Minimum width to match beam flange. Preferred alternative: extend plate flush with column flanges. 

7. Remove weld tabs to 1⁄4” maximum from edge of continuity plate. Grind end of weld smooth (250 min), not flush. Do not gouge column flange. 

8. Beam connection, see individual prequalifications. 

Figure 6-9 Typical Continuity and Doubler Plates 

6.4.3.2 Panel Zone Strength 

Moment-resisting connections should be proportioned either so that shear yielding of the 

panel zone initiates at the same time as flexural yielding of the beam elements, or so that all 

yielding occurs in the beam. The following procedure is recommended: 

Step 1: Calculate t, the thickness of the panel zone that results in simultaneous yielding of the 

panel zone and beam from the following relationship: 

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h d - CM b 

y c 

t = 

h 

(0.9) 0.55F Rycdc (db - t yc fb ) (6-10)

where: 

h is the average story height of the column, measured from the midpoint of the 

column above the beam to the midpoint of the column below the beam. 

Ryc is the ratio of the expected yield strength of the column material to the minimum 

specified yield strength, in accordance with the 1997 AISC Seismic Provisions. 

Mc and Cy are the coefficients defined in Section 6.3.7 and Section 6.3.8 of these 

Recommended Criteria, respectively, and other terms are as defined in the AISC-LRFD 

Specifications. 

Step 2: If t, as calculated, is greater than the thickness of the column web, provide doubler 

plates, or increase the column size to a section with adequate web thickness. 

Where doubler plates are required, the thickness should be determined as described above, 

and they should be proportioned and welded as described in the 1997 AISC Seismic Provisions. 

QC/QA Category BL/L procedures are defined in FEMA-353. 

For connections designed using project-specific qualifications, the panel zone strength should 

match that of the tested connections. 

Commentary: Several aspects of the methodology for the design of panel zones, 

as contained in the 1997 AISC Seismic Provisions, are considered to require 

revision, based on studies conducted by this project. As described in FEMA355D, the best performance is likely to be achieved when there is a balance of 

beam bending and panel zone distortion. The equations given are intended to 

provide panel zones that are just at the onset of yielding at the time the beam 

flange begins to yield. 

The procedure recommended in this design criteria varies significantly from 

that contained in the 1997 AISC Seismic Provisions, but the results are not 

dramatically different. For most column sizes results will be similar to methods 

used in the past. For columns with thick flanges, the methods herein will result in 

the need for moderately thicker panel zones than in the past. 

6.4.3.3 Connections to Column Minor Axis 

Connections to the minor axis of a column should be qualified by testing following the 

procedures of Section 6.9. If minor-axis connections are to be used in conjunction with majoraxis connections to the same column, the testing program should include biaxial bending effects 

at the connection. 

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Commentary: In general, the prequalified connections have not been tested for 

use with columns oriented so that beams connect to the minor axis of the column. 

Two tests of Reduced Beam Section connections in this orientation were 

conducted, and indicated good performance. These tests were conducted to 

provide a general indication of the possible performance of weak axis 

connections, but are not considered to comprise a sufficient database for 

prequalification of such connections. 

6.4.3.4 Attachment of Other Construction 

Welded or bolted attachment for exterior facades, partitions, ductwork, piping, or other 

construction should not be placed in the hinging area of moment frame beams. The hinging area 

is defined as one half of the beam depth on either side of the theoretical hinge point as described 

in the prequalification data table for each connection detail. It is recommended that bolt holes for 

this type of construction not be permitted between the face of the column and six inches, 

minimum, beyond the extreme end of the hinging area. Outside the described area, a calculation 

should be made to ensure sufficient net section to avoid fracture, based on moments calculated 

using the expected moment at the hinge point. Welding between the column face and the near 

edge of the hinging area should be carefully controlled to avoid creation of stress concentrations 

and application of excessive heat. Specifications and drawings should clearly indicate that 

anchorage shall not be made in the areas described and this should be coordinated with the 

architect and other members of the design team. 

Commentary: It is common for precast panels and other facade elements, as well 

as other construction, to be anchored to members of the steel frame through the 

use of welds, bolts, powder-driven fasteners, or other fasteners. Such anchorage 

is often not considered by the engineer and is not performed with the same care 

and quality control as afforded the main building structure. Such anchorage, 

when made in an area of high stress, can lead to stress concentrations and 

potential fracture. 

6.4.4 Bolted Joint Requirements 

6.4.4.1 Existing Conditions 

When evaluating existing structures, the condition of bolted connections should be 

determined based on the AISC and Research Council on Structural Connections (RCSC) 

specifications appropriate to the design and construction years, and on the following criteria: 

• Representative samples of bolts should be inspected to determine markings and 

classifications. Where bolts cannot be properly identified visually, representative samples 

should be removed and tested to determine tensile strength in accordance with ASTM F606 

and the bolt classified accordingly. Alternatively, bolts may be assumed to be A307. 

• Any evidence of yielding in the connection plates indicates that the high-strength bolts are 

effectively in the snug-tight condition regardless of their original installation condition. If 

bolts have been identified as ASTM A325 and are not in a snug-tight condition they should 

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be re-tightened or replaced. If bolts have been identified as ASTM A490 and are not in a 

snug-tight condition, they should be replaced. Re-tightening or installation of bolts should be 

to a pretensioned condition in accordance with the 1997 AISC or 1996 RCSC criteria. 

6.4.4.2 Connection Upgrades 

When upgrading existing connections, the capacity of bolted elements of the connection shall 

be determined based on the AISC and RCSC specifications appropriate to the design and 

construction years, and the following criteria: 

• Bolts intended to transfer load in the shear/bearing mode should be installed according to the 

slip critical criteria. 

• Bolts intended to transfer load by tension should be pre-tensioned. 

• Bolts intended for use in proprietary connections, such as a viscous damping system, should 

be installed using the instructions applicable to the test data for the system. 

• Bolted joints should not be upgraded by sharing loads with weld reinforcement. Any welded 

reinforcement shall be designed to transfer all the load, independent of the bolt capacity. 

6.5 Prequalified Connection Details – General 

Prequalified connection and connection upgrade details are permitted to be used for moment 

frame connections for the types of moment frames and ranges of the various design parameters 

indicated in each prequalification description. Project-specific testing should be performed to 

demonstrate the adequacy of connection and upgrade details that are not listed herein as 

prequalified, or are used outside the range of parameters indicated in the prequalification. 

Designers should follow the procedures outlined in Section 6.9 for use of nonprequalified 

connection and upgrade details. 

Commentary: The following criteria were applied to connection and upgrade 

details listed as prequalified: 

1. There is sufficient experimental and analytical data on the connection 

performance to establish the likely yield mechanisms and failure modes for 

the connection. 

2. Rational models for predicting the resistance associated with each 

mechanism and failure mode have been developed. 

3. Given the material properties and geometry of the connection, a rational 

procedure can be used to estimate which mode and mechanism controls the 

behavior and the deformation capacity (that is, the drift angle) that can be 

attained from the controlling conditions. 

4. Given the models and procedures, the existing data base is adequate to 

permit assessment of the statistical reliability of the connection. 

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Some of the connection and upgrade details in the following sections are only 

prequalified for use in Ordinary Moment Frames (OMFs), while others are 

prequalified for both OMF and Special Moment Frame (SMF) use. In general, 

when a connection is qualified for use in SMF systems, it is also qualified for use 

in OMF systems, with fewer restrictions on size, span, and other parameters than 

are applied to the SMF usage. Very little extrapolation has been applied in the 

prequalification limitations for SMFs, while some judgement has been applied to 

permit extrapolation for OMFs, based on the significantly lower rotational 

demands applicable to those systems. 

6.5.1 Load Combinations and Resistance Factors 

Design procedures for prequalified connection upgrades contained in Section 6.6 are 

formatted on an expected strength basis, as opposed to either a Load and Resistance Factor 

Design basis or Allowable Stress Design basis. Loading used in these design formulations is 

generally calculated on the basis of the stresses induced in the assembly at anticipated yielding of 

the beam-column connection assembly. Where these design procedures require that earthquake 

loading be applied simultaneously with dead and live loading, the applicable load combinations 

of the 1997 AISC Seismic Provisions apply. Resistance factors should not be applied except as 

specifically required by the individual design procedure. 

6.6 Prequalified Connection Upgrades 

This section provides prequalification data for various alternative types of welded steel 

moment-frame (WSMF) connection upgrade details. Table 6-6 lists the various alternative 

connection upgrade details that have been prequalified, together with the structural system (SMF 

or OMF) for which they are prequalified for use in Simplified Upgrade, and reference to the 

section of these Recommended Criteria where detailed information may be found. Refer to these 

individual reference sections for specific limits on the applicability of the prequalification, for 

specific performance data for use with Systematic Upgrade and for specific design procedures 

and details. 

Table 6-6 Prequalified Welded Fully Restrained Connection Upgrade Details 

Connection Type Criteria 

Section 

Structural System 

Improved welded unreinforced flange IWURF 6.6.1 OMF 

Welded bottom haunch WBH 6.6.2 OMF, SMF 

Welded top and bottom haunch WTBH 6.6.3 OMF, SMF 

Welded cover plated flange WCPF 6.6.4 OMF, SMF 

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Commentary: FEMA-355D – State of the Art Report on Connection 

Performance, provides extensive information on the testing and performance of 

these connections that is not repeated in this document. The data presented in 

FEMA-355D have been used in support of development of the prequalification 

performance data, design procedures, and limitations on design parameters for 

these connections presented herein. 

6.6.1 Improved Welded Unreinforced Flange (IWURF) Connection 

This section provides recommended criteria for design of connection upgrades intended to 

improve existing unreinforced, welded flange connections by improving the existing welded 

joints in the connection. This connection upgrade is prequalified only for Ordinary Moment 

Frame applications. Upgrade is accomplished through replacement of existing complete joint 

penetration groove welds of low-notch-toughness material and potentially having significant root 

defects, with new welds conforming to current construction requirements for welded steel 

moment-frame construction as shown in Figures 6-10 and 6-11. In addition, other elements of 

the connection, including panel zones and column flanges are reinforced, as required, to conform 

to the general recommendations of Section 6.4. Table 6-7 tabulates the limits of applicability of 

this prequalified connection upgrade and associated performance qualification data. 

Commentary: This connection upgrades the typical pre-Northridge “prescriptive 

connection” commonly in use prior to the 1994 Northridge earthquake. After 

significant study, it has been concluded that with several improvements this 

connection can be made to perform reliably in frames designed as Ordinary 

Moment Frames as long as beam sizes are limited as indicated in Table 6-7. 

The improvements required for this connection include the following: 

1. Removal of existing low-toughness weld metal and replacement with weld 

metal with appropriate toughness; 

2. Removal of bottom flange weld backing, back-gouging and addition of a 

reinforcing weld; 

3. Removal of weld tabs; 

4. Improvements to weld quality control and quality assurance requirements and 

methods. 

For best performance of this connection type some limited panel zone yielding 

is beneficial. For this reason, it is recommended that panel zones not be overreinforced. 

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Notes: 

1. See Figure 6-11 for welding requirements at these locations. 

2. Existing bolted shear tab. 

3. Existing or added continuity plates and web doubler plate. See Figure 6-9. 

Figure 6-10 Improved Welded Unreinforced Flange Connection 

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Notes: 

1. Gouge out existing weld at both the top and bottom flange and prepare joints for new weld. 

2. Complete joint penetration groove weld at top and bottom flanges. At top flange, either 

(A), remove weld backing, backgouge, and add 5/16” minimum fillet weld, or (B), leave 

backing in place and add 5/16” fillet under backing. At bottom flange, remove weld 

backing, backgouge, and add 5/16” minimum fillet weld. Weld is QC/QA Category AH/T. 

3. Existing weld access hole to remain unmodified. 

Figure 6-11 Welding Requirements at Improved Welded Unreinforced Flange 

Connection 

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Table 6-7 Prequalification Data for Improved Welded Unreinforced Flange Connections 

Applicability Limits 

General: 

Applicable systems OMF 

Hinge location distance sh dc / 2 + db / 2 

Critical Beam Parameters: 

Depth W36 and shallower 

Minimum span-to-depth ratio 7 

Flange thickness 1” maximum 

Permissible material specifications A7, A36, A572 Gr. 50 

Critical Column Parameters: 

Depth Not limited 

Permissible material specifications A7, A36, A572 Gr. 50 

Beam/Column Relations: 

Panel zone strength Section 6.4.3.2, Cpr = 1.1 

Column/beam bending strength No requirement (OMF) 

Connection Details: 

Web connection Existing bolted shear tab 

Continuity plate thickness Section 6.4.3.1 

Flange welds Figures 6-10 and 6-11 

Weld electrodes Sections 6.4.2.4 and 6.4.2.5 

Weld access holes Existing weld access hole 

Performance Data: 

Strength degradation rotation - qSD, radians 0.031 - 0.0003db 

Immediate Occupancy rotation - qIO, radians 0.015, but not greater than qSD 

Resistance factor, Immediate Occupancy, f 0.9 

Collapse Prevention drift angle - qU, radians 0.060 - 0.0006db 

Resistance factor, Collapse Prevention, f 0.9 

Notes: db= beam depth, inches; dc = column depth, inches. 

6.6.1.1 Design Procedure 

Step 1: Calculate Mpr, at hinge location, sh, according to methods of Section 6.3.5. 

Step 2: Calculate Vp, at hinge location, sh, according to methods of Section 6.3.6. 

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Step 3: Calculate Mc, Mf, and Cy as described in Section 6.3.7 and 6.3.8. 

Step 4: Calculate the required panel zone thickness using the procedures of Section 6.4.3.2. 

Step 5: Check requirements for Continuity Plates according to Section 6.4.3.1. 

Step 6: Detail the connection as shown in Figure 6-10 and 6-11. 

Commentary: There is more research information available on unreinforced 

beam-to-column connections than there is on any other type of steel momentframe connection. Not only were these connections extensively studied prior to 

the 1994 Northridge earthquake, they have been even more extensively studied in 

the aftermath. Many of the studies focused on the connection as used in pre-1994 

practice, with bolted web connection, and flange welds with unrated or low notch 

toughness and with backing left in place, while other studies have been focused on 

improvements to the connection, including those improvements recommended in 

this section. 

These tests give widely scattered results, but in general, indicate that when 

weld metal with sufficient notch toughness is used and workmanship is 

maintained at an appropriate level, these connections can reliably perform 

adequately for service in Ordinary Moment Frame, if not Special Moment Frame 

systems. Additional information may be found in FEMA-355D, State of the Art 

Report on Connection Performance. 

6.6.2 Welded Bottom Haunch (WBH) Connection 

This connection upgrade is accomplished by converting the existing welded unreinforced 

(WUF) connection into a haunched connection, with a single haunch present at the bottom beam 

flange. This connection upgrade is prequalified for both OMF and SMF applications. If the weld 

of the top beam flange to the column is made with weld metal with low or unclassified notch 

toughness, then, in addition to welding the new haunch at the bottom beam flange, this top beam 

flange weld must be gouged out and replaced with weld metal conforming to the 

recommendations of Sections 6.4.2.3 and 6.4.2.4 to obtain SMF service. The general 

requirements of Section 6.4 should be complied with. Figure 6-12 provides a typical detail for 

this connection. Table 6-8 presents performance qualification data for the connection. Refer to 

AISC Steel Design Guide Series 12 (Gross et al., 1999) for supplemental information to the 

design procedure given in Section 6.6.2.1. 

6.6.2.1 Design Procedure 

Step 1: Calculate Mpr, at hinge location, sh, according to methods of Section 6.3.5. 

Step 2: Calculate Vp, at hinge location, sh, according to methods of Section 6.3.6. 

Step 3: Calculate Mc, Mf, and Cy as described in Section 6.3.7 and 6.3.8. 

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Step 4: Calculate the required panel zone thickness using the procedures of Section 6.4.3.2. 

Step 5: Check requirements for Continuity Plates according to Section 6.4.3.1. 

Step 6: Size the haunch according to the criteria outlined in AISC Steel Design Guide 

Series 12. 

Step 7: Detail the connection as shown in Figure 6-12. 

Notes 

1. For OMF connection, existing weld can remain. For SMF connection, see Figure 6-11. 

2. Existing bolted shear tab. 

3. Existing continuity plates and web doubler plate. See Figure 6-9. 

4. WT haunch. 

5. New 1⁄2”-minimum stiffener plates each side. 

6. Haunch welds, see Sections 6.4.2.3 and 6.4.2.4, QC/QA category AH/T. 

7. Stiffener CJP welds; see Sections 6.4.2.3 and 6.4.2.4, QC/QA Category BM/T. 

8. Stiffener fillet welds, 5/16” minimum. QC/QA Category CL/L. 

Figure 6-12 Welded Bottom Haunch (WBH) Connection 

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Table 6-8 Prequalification Data for Welded Bottom Haunch (WBH) Connection 

Applicability Limits 

General: 

Applicable systems OMF, SMF 

Hinge location distance sh dc /2 + lh from center of column 

Critical Beam Parameters: 

Depth range Up to W36 

Minimum span-to-depth ratio OMF: 5 

SMF: 7 

Flange thickness OMF: 1-1/2” maximum 

SMF: 1” maximum 

Permissible material specifications A7, A36, A572 Gr. 50 

Beam flange welds OMF: Existing welds can remain. 

SMF: Sections 6.4.2.3 and 6.4.2.4 

Critical Column Parameters: 

Depth OMF: Not limited 

SMF: W12, W14 

Permissible material specifications A7, A36, A572 Gr. 50 

Beam / Column Relations: 

Panel zone strength OMF: Section 6.4.3.2, Cpr = 1.1 

SMF: Section 6.4.3.2 

Column/beam bending strength ratio OMF: No requirement 

SMF: Section 6.4.1.1 

Connection Details: 

Web connection Existing bolted shear tab 

Continuity plate thickness At beam flanges: Section 6.4.3.1 

At haunch: match haunch width and thickness 

Haunch welds Sections 6.4.2.3 and 6.4.2.4 

Details of Haunch Design: 

Haunch size and strength criteria Haunch to be sized by criteria as outlined in AISC Steel 

Design Guide Series 12 (Gross et al., 1999) 

Performance Data: 

Strength degradation rotation - qSD, radians 0.038 

Immediate Occupancy rotation - qIO, radians 0.020 

Resistance factor, Immediate Occupancy, f 0.9 

Collapse Prevention drift angle - qU – radians 0.06 

Resistance factor, Collapse Prevention, f 0.9 

Note: dc = column depth 

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6.6.3 Welded Top and Bottom Haunch (WTBH) Connection 

This connection upgrade is accomplished by attaching a new welded haunch to both the top 

and bottom flanges of the existing beam connection. This connection upgrade is prequalified for 

both OMF and SMF applications. Existing welds in the connection need not be gouged out, nor 

replaced, for OMF applications. For SMF applications, in addition to installing the new 

haunches, if the beam flange welds to the column are made with weld metal of unclassified or 

low notch toughness, these welds must be gouged out and replaced with weld metal conforming 

to the recommendations of Sections 6.4.2.3 and 6.4.2.4. Design is accomplished to 

accommodate the general requirements of Section 6.4. Figure 6-13 shows a typical detail for this 

connection. Table 6-9 provides performance qualification data. 

Notes 

1. For OMF connection, weld can remain. For SMF connection, see Figure 6-11. 

2. Existing bolted shear tab. 

3. Existing continuity plates and web doubler plate. See Figure 6-9. 

4. WT haunches. 

5. New 1⁄2"-minimum stiffener plate each side. 

6. Haunch welds, see Sections 6.4.2.3 and 6.4.2.4, QC/QA category AH/T. 

7. Stiffener CJP welds; see Sections 6.4.2.3 and 6.4.2.4, QC/QA Category BM/T. 

8. Stiffener fillet welds, 5/16” minimum. QC/QA Category CL/L. 

Figure 6-13 Welded Top and Bottom Haunch (WTBH) Connection 

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Table 6-9 Prequalification Data for Welded Top and Bottom Haunch 

(WTBH) Connections 

Applicability Limits 

General: 

Applicable systems OMF, SMF 

Hinge location distance sh dc /2 + lh from center of column 

Critical Beam Parameters: 

Depth range Up to W36 

Minimum span-to-depth ratio OMF: 5 

SMF: 7 

Flange thickness OMF: 1-1/2” maximum 

SMF: 1” maximum 

Permissible material specifications A7, A36, A572 Gr. 50 

Beam flange welds OMF: Existing welds can remain. 

SMF: Sections 6.4.2.3 and 6.4.2.4 

Critical Column Parameters: 

Depth OMF: Not limited 

SMF: W12, W14 

Permissible material specifications A7, A36, A572 Gr. 50 

Beam / Column Relations: 

Panel zone strength OMF: Section 6.4.3.2, Cpr = 1.1 

SMF: Section 6.4.3.2 

Column/beam bending strength ratio OMF: No requirement 

SMF: Section 6.4.1.1 

Connection Details: 

Web connection Existing bolted shear tab 

Continuity plate thickness At beam flanges: Section 6.4.3.1 

At haunch: match haunch width and thickness 

Haunch welds Section 6.4.2.3 and 6.4.2.4 

Details of Haunch Design: 

Haunch size and strength criteria Haunch to be sized by criteria as outlined in AISC Steel 

Design Guide Series 12 (Gross et al., 1999) 

Performance Data: 

Strength degradation rotation - qSD, radians 0.038 

Immediate Occupancy rotation - qIO, radians 0.02 

Resistance factor, Immediate Occupancy, f 0.9 

Collapse Prevention drift angle - qU – radians 0.058 

Resistance factor, Collapse Prevention, f 0.9 

Note: dc = depth of column, inches 

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6.6.3.1 Design Procedure 

Step 1: Calculate Mpr, at hinge location, sh, according to methods of Section 6.3.5. 

Step 2: Calculate Vp, at hinge location, sh, according to methods of Section 6.3.6. 

Step 3: Calculate Mc, Mf, and Cy as described in Section 6.3.7 and 6.3.8. 

Step 4: Calculate the required panel zone thickness using the procedures of Section 6.4.3.2. 

Step 5: Check requirements for Continuity Plates according to Section 6.4.3.1. 

Step 6: Size the haunches according to the criteria outlined in AISC Steel Design Guide 

Series 12 (Gross, et al., 1999). 

Step 7: Detail the connection as shown in Figure 6-13. 

6.6.4 Welded Cover Plated Flange (WCPF) Connection 

This connection upgrade is accomplished by attaching new cover plates to both the top and 

bottom flanges of the existing beam. This connection upgrade is prequalified for both OMF and 

SMF applications. Existing welds in the connection need not be gouged out, nor replaced, for 

OMF applications. In addition to welding the new cover plates, if the beam flange welds to the 

column are made with welds having notch toughness that is either not classified or low, this weld 

must be gouged out and replaced with weld metal conforming to the recommendations of 

Sections 6.4.2.3 and 6.4.2.4 to obtain SMF service. Design is accomplished to accommodate the 

general requirements of Section 6.4. Figure 6-14 shows a typical detail for this connection. 

Table 6-10 provides prequalification limitations. 

6.6.4.1 Design Procedure 

Step 1: Calculate Mpr, at hinge location, sh, according to methods of Section 6.3.5. 

Step 2: Calculate Vp, at hinge location, sh, according to methods of Section 6.3.6. 

Step 3: Calculate Mc, Mf, and Cy as described in Section 6.3.7 and 6.3.8. 

Step 4: Calculate the required panel zone thickness using the procedures of Section 6.4.3.2. 

Step 5: Check requirements for Continuity Plates according to Section 6.4.3.1. 

Step 6: Size the cover plates. When cover plates are to be field welded, the top cover plate 

should be narrower than the beam flange and the bottom cover plate should be wider. 

The area of the cover plates should be sized to satisfy the following relationship: 

(kZb + Acp (db + tcp ))Fy ‡ M f (6-11) 

where: 

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k = 0.4 for OMF and 1.0 for SMF connections

Acp = cross-section area of the cover plate, square inches

db = depth of the beam, inches

tcp = thickness of the cover plate, inches

The remainder of the terms are as defined in Section 6.3 and 6.4. 

Step 7: Detail the connection as shown in Figure 6-14. 

Notes: 

1. For OMF connection, weld can remain. For SMF connection, see Figure 6-11. 

2. Existing bolted shear tab. 

3. Existing continuity plates and web doubler plate. See Figure 6-8. 

4. Cover plates. 

5. Cover plate CJP welds, see Section 6.4.2.3 and 6.4.2.4, QC/QA Category AH/T. 

6. Cover plate fillet welds, QC/QA Category BH/L. 

Figure 6-14 Welded Cover Plated Flange (WCPF) Connection 

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Table 6-10 Prequalification Data for Welded Cover Plated Flange Connections 

Applicability Limits 

General: 

Applicable systems OMF, SMF 

Hinge location distance sh dc /2 + lcp from center of column 

Critical Beam Parameters: 

Depth range Up to W36 

Minimum span-to-depth ratio OMF: 5 

SMF: 7 

Flange thickness OMF: 1-1/2: maximum 

SMF: 1” maximum 

Permissible material specifications A7, A36, A572 Gr. 50 

Beam flange welds OMF: Existing welds can remain. 

SMF: Sections 6.4.2.3 and 6.4.2.4. 

Critical Column Parameters: 

Depth OMF: Not limited 

SMF: W12, W14 

Permissible material specifications A7, A36, A572 Gr. 50 

Beam / Column Relations: 

Panel zone strength OMF: Section 6.4.3.2, Cpr = 1.1 

SMF: Section 6.4.3.2 

Column/beam bending strength ratio OMF: No requirement 

SMF: Section 6.4.1.1 

Connection Details: 

Relative size and proportions of cover plate Section 6.6.4.1, Step 6. 

Web connection Existing bolted shear tab. 

Continuity plate thickness Section 6.4.3.1 

Cover plate welds Section 6.4.2.3 and 6.4.2.4 

Performance Data: 

Strength degradation rotation - qSD, radians 0.066 - 0.0011db 

Immediate Occupancy rotation - qIO, radians 0.02, but not greater than qSD 

Resistance factor, Immediate Occupancy, f 0.9 

Collapse Prevention drift angle - qU, radians 0.066 - 0.0011db 

Resistance factor, Collapse Prevention, f 0.9 

Notes: db= beam depth, inches, dc= column depth 

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6.7 New Moment Frames and Moment-Resisting Connections 

In some cases, it may be desirable to upgrade an existing steel moment-frame building by 

introducing new steel moment frames. This can be accomplished either with the addition of new 

framing, or the modification of existing framing not originally intended to participate in lateral 

resistance. New moment-resisting connections, introduced for such purpose, should be designed 

in accordance with the design procedures presented in FEMA-350, Recommended Seismic Design 

Criteria for New Steel Moment-Frame Buildings, and constructed in accordance with FEMA353, Recommended Specifications and Quality Assurance Guidelines for Steel Moment-Frame 

Construction for Seismic Applications. Table 6-11 presents performance data for connections 

that have been prequalified for use in new construction. The table may be used in assessing the 

effectiveness of new or modified framing employing these connections to achieve desired 

performance goals. 

Commentary: Upgrade of existing WSMF buildings with the addition of new steel 

moment frames, or the modification of existing gravity frames to provide lateral 

resistance, will typically not be an effective upgrade strategy. This is because 

steel moment frames are inherently flexible and it is unlikely that the addition of 

new frames, by themselves, will be sufficient to control building drifts to levels 

that will protect existing WSMF connections from damage. 

6.8 Proprietary Connections 

This section presents information on several types of fully restrained connection technologies 

that have been developed on a proprietary basis. These connection technologies are not 

categorized in these Recommended Criteria as prequalified, as the SAC Joint Venture has not 

examined the available supporting data in sufficient detail to confirm that they meet appropriate 

prequalification criteria. However, these proprietary connections have been evaluated by some 

enforcement agencies and found to be acceptable for specific projects and in some cases for 

general application within the jurisdiction’s authority. Use of these technologies without the 

express permission of the licensor may be a violation of intellectual property rights, under the 

laws of the United States. 

Discussion of several types of proprietary connections are included herein. Other proprietary 

connections may also exist. Inclusion or exclusion of proprietary connections in these 

Recommended Criteria should not be interpreted as either an approval or disapproval of these 

systems. The descriptions of these connections contained herein have in each case been prepared 

by the developer or licensor of the technology. This information has been printed with their 

permission. Neither the Federal Emergency Management Agency nor the SAC Joint Venture 

endorses any of the information provided or any of the claims made with regard to the attributes 

of these technologies or their suitability for application to specific projects. Designers wishing to 

consider specific proprietary connections for use in their structures should consult both the 

licensor of the connection and the applicable enforcement agency to determine the applicability 

and acceptability of the individual connection for the specific design application. 

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Table 6-11 Performance Data for Prequalified Moment-Resisting Connections for New 

Framing 

Strength 

Degradation1 

Immediate 

Occupancy 

Collapse 

Prevention1 

Connection Type qSD qIO 2 f qU f 

Welded Unreinforced Flange, 

Bolted Web 

(WUF-B) 

0.031-0.0003db 0.020 0.9 0.060-0.0006db 0.9 

Welded Unreinforced Flange, 

Welded Web 

(WUF-W) 

0.051 0.020 0.9 0.064 0.9 

Free Flange 

(FF) 

0.077-0.0012db 0.020 0.9 0.104-0.0016db 0.9 

Reduced Beam Section 

(RBS) 

0.060-0.0003db 0.020 0.9 0.080-0.0003db 0.9 

Welded Flange Plate 

(WFP) 

0.04 0.020 0.9 0.07 0.9 

Bolted Unstiffened End Plate 

(BUEP) 

0.071-0.0013db 0.020 0.9 0.081-0.0013db 0.9 

Bolted Stiffened End Plate 

(BSEP) 

0.071-0.0013db 0.020 0.9 0.081-0.0013db 0.9 

Bolted Flange Plate 

(BFP) 

0.12-0.0023db 0.020 0.9 0.10-0.0011db 0.9 

Double Split Tee 

(DS) 

0.12-0.0032db 0.020 0.9 0.14-0.0032db 0.9 

Notes: 

Values in this table apply only to connections and framing that comply in all respects with the 

prequalification limits indicated in FEMA-350, Recommended Seismic Design Criteria for New Steel 

Moment-Frame Buildings and FEMA-353, Recommended Specifications and Quality Assurance 

Guidelines for Steel Moment-Frame Construction for Seismic Applications. 

1. For connections that are prequalified in FEMA-350 for either SMF or OMF service, the values 

indicated apply for framing and connections that comply with the applicability limits for SMF 

service. When framing and connections comply with the applicability limits for OMF service 

but not for SMF service, 1⁄2 the tabulated values shall be used. 

2. The value of qIO shall not be taken greater than the value for qSD. 

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6.8.1 Side Plate (SP) Connection 

The proprietary Side Plate connection system is a patented technology shown schematically 

in Figure 6-15 for its application to upgrade of existing construction. Physical separation 

between the face of the column flange and the end of the beam eliminates peaked triaxial stress 

concentrations. Physical separation is achieved by means of parallel full-depth side plates that 

eliminate reliance on through-thickness properties and act as discrete continuity elements to 

sandwich and connect the beam and the column. The increased stiffness of the side plates 

inherently stiffens the global frame structure and eliminates reliance on panel zone deformation 

by providing three panel zones [i.e., the two side plates plus the column’s own web]. Top and 

bottom beam flange cover plates are used, when dimensionally necessary, to bridge the 

difference between the flange widths of the beam and the column. 

This connection system uses all fillet-welded fabrication. All fillet welds are made in either 

the flat or horizontal position using column tree construction. For new construction, shop 

fabricated column trees and link beams are erected and joined in the field using one of four link 

beam splice options to complete the moment-resisting frame. Link beam splice options include a 

fully welded CJP butt joint, bolted matching end plates, fillet-welded flange plates, and bolted 

flange plates. 

Figure 6-15 Proprietary Side Plate Connection – Application to Existing Construction 

All connection fillet welds are loaded principally in shear along their length. Moment 

transfer from the beam to the side plates, and from the side plates to the column, is accomplished 

with plates and fillet welds using equivalent force couples. Beam shear transfer from the beam’s 

web to the side plates is achieved with vertical shear plates and fillet welds. The side plates are 

designed with adequate strength and stiffness to force all significant plastic behavior of the 

connection system into the beam, in the form of flange and web local buckling centered at a 

distance of approximately 1/3 the depth of the beam away from the edge of the side plates. 

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All full-scale cyclic testing of this connection system was conducted at the Charles Lee 

Powell Structural Research Laboratories, University of California, San Diego, under the direction 

of Professor Chia-Ming Uang. Testing included both prototype uniaxial and biaxial dual strong 

axis tests. Independent corroborative nonlinear analyses were conducted by the University of 

Utah and by Myers, Houghton & Partners, Structural Engineers. 

Independent prequalification of this connection system was determined by ICBO Evaluation 

Service, Inc., in accordance with ICBO ES Acceptance Criteria for Qualification of Steel 

Moment-Frame Connection Systems (AC 129-R1-0797), and was corroborated by the City of Los 

Angeles Engineering Research Section, Department of Building and Safety. These invoke the 

qualification procedures contained in FEMA 267/267A/267B; AISC Seismic Provisions for 

Structural Steel Buildings, dated April 15, 1997; and County of Los Angeles Current Position on 

Design and Construction of Welded Moment-Resisting Frame Systems CP-2, dated August 14, 

1996. Refer to ICBO Evaluation Service, Inc., Evaluation Report No. 5366, issued January 1, 

1999, and to City of Los Angeles Research Report: COLA RR 25393 for allowable values and 

conditions of use. Additional independent jurisdictional scrutiny of this connection system, by 

Karl H. Frank, Ph.D., Egor P. Popov, Ph.D., C. Mark Saunders, S.E., and Robert L. Schwein, 

P.E. is contained in the Los Angeles County Technical Advisory Panel (LACO-TAP) SMRF 

Bulletin No. 3 on Steel Moment-Resisting Frame Connection Systems, County of Los Angeles, 

Department of Public Works, dated March 4, 1997. Additional design information for this 

connection type may be obtained from the licensor. 

The Side Plate connection for upgrade construction differs from its configuration for new 

construction by featuring an initial opening in each side plate to permit welding access, saving 

the cut-out pieces of plate for use as closure plates to close the access window after welding is 

completed. All new welds are fillet welds loaded principally in shear along their length. The 

existing Complete Joint Penetration (CJP) welds joining the beam flanges to the column flange 

are removed by airarcing to eliminate reliance on through-thickness properties and triaxial stress 

concentrations. The existing shear tab of the steel moment-frame beam(s) is left in place to 

provide gravity support. Existing continuity plates may be left in place to act as horizontal shear 

plates as depicted in Figure 6-15. 

6.8.2 Slotted Web (SW) Connection 

This proprietary connection (Seismic Structural Design Associates, Inc. US Patent No. 

5,680,738 issued 28 October 1997) is shown schematically in Figure 6-16. It is similar to the 

popular field welded–field bolted beam-to-column moment frame connection, shown in the 

current AISC LRFD and ASD steel design manuals, that has become known as the “preNorthridge” connection. Based upon surveys of seismic connection damage, modes of fracture, 

reviews of historic tests, and recent ATC-24 protocol tests, it was concluded by SEAOC (1996 

Blue Book Commentary) that the pre-Northridge connection is fundamentally flawed and should 

not be used in the new construction of seismic moment frames. Subsequent finite element 

analyses and strain gage data from ATC-24 tests of this pre-Northridge connection have shown 

large stress and strain gradients horizontally across and vertically through the beam flanges and 

welds at the face of the column. These stress gradients produce a prying moment in the beam 

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flanges at the weld access holes and in the flange welds at the column face that lead to beam 

flange and weld fractures and column flange divot modes of connection fracture. Moreover, 

these same studies have also shown that a large component, typically 50%, of the vertical beam 

shear and all of the beam moment, is carried by the beam flanges/welds in the pre-Northridge 

connection. 

However, by (1) separating the beam flanges from the beam web in the region of the 

connection and (2) welding the beam web to the column flange, the force, stress and strain 

distributions in this field welded-field bolted connection are changed dramatically in the 

following ways: 

1. The vertical beam shear in the beam flanges/welds is reduced from typically 50% to typically 

3% so that essentially all vertical shear is transferred to the column through the beam web 

and shear plate. 

2. Since most W sections have a flange to beam modulus ratio of 0.65 < Zflg /Z < 0.75, both the 

beam web and flange separation and the beam web to column flange weldment force the 

beam web to resist its portion of the total beam moment. 

3. The beam web separation from the beam flange reduces the large stress and strain gradients 

across and through the beam flanges by permitting the flanges to flex out of plane. Typically, 

the elastic stress and strain concentration factors (SCFs) are reduced from 4.0 to 5.0 down to 

1.2 to 1.4, which dramatically reduces the beam flange prying moment and the accumulated 

plastic strain and ductility demand under cyclic loading. These attributes enhance and extend 

the fatigue life of this moment frame connection. 

4. The lateral-torsional mode of beam buckling that is characteristic of non-slotted beams is 

circumvented. The separation of the beam flanges and beam web allow the flanges and web 

to buckle independently and concurrently, which eliminates the twisting mode of buckling 

and its associated torsional beam flange/weld stresses. Elimination of this buckling mode is 

particularly important when the exterior cladding of the building is supported by seismic 

moment frames that are located on the perimeter of the building. 

5. Residual weldment stresses are significantly reduced. The separation of the beam web and 

flanges in the region of the connection provides a long structural separation between the 

vertical web and horizontal flange weldments. 

The slotted web (SW) connection design rationale that sizes the beam/web separation length, 

shear plate and connection weldments, is based upon ATC-24 protocol test results and inelastic 

finite element analyses of the stress and strain distributions and buckling modes. Incorporated in 

this rationale are the UBC and AISC Load and Resistance Factor Design (LRFD) Specifications 

and the AISC Seismic Design Provisions for Steel Buildings. 

Seismic Structural Design Associates (SSDA) has successfully completed ATC-24 protocol 

tests on beams ranging from W27x94 to W36x280 using columns ranging from W14x176 to 

W14x550. None of these assemblies experienced the lateral-torsional mode of buckling that is 

typical of non-slotted beam and column assemblies. 

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Figure 6-16 Proprietary Slotted Web Connection 

Both analytical studies and ATC-24 protocol tests have demonstrated that the Seismic 

Structural Design Associates (SSDA) Slotted Web connection designs develop the full plastic 

moment capacity of the beam and do not reduce the elastic stiffness of the beam. All of the above 

attributes of this proprietary connection enhance its strength and ductility, which makes it 

applicable for use in retrofit of existing seismic moment frames. Specific qualification and 

design information for the Slotted Web connection may be obtained from the licensor. 

6.8.3 Bolted Bracket (BB) Connection 

This connection type is shown schematically in Figure 6-17. Beam shear and flexural 

stresses are transferred to the column through a pair of heavy bolted brackets, located at the top 

and bottom beam flanges. The concept of using bolted brackets to connect beams to columns 

rigidly is within the public domain, but generic prequalification data have not been developed for 

this connection. One licensor has developed patented steel castings of the bolted brackets, for 

which specific design qualification data has been prepared. Specific qualification and design 

information for this connection may be obtained from the licensor. 

6.9 Project-Specific Testing of Nonprequalified Connections 

This section provides recommended criteria for design and project-specific qualification of 

connections and connection upgrades for which there is no current prequalification. 

Recommended criteria are also provided for prequalified details which are to be utilized outside 

the parametric limitations for a current prequalification. Project-specific qualification includes a 

program of connection assembly prototype testing, supplemented by a suitable analytical 

procedure that permits prediction of behavior identified in the testing program. 

Commentary: While it is not the intent of these Recommended Criteria to require 

testing for most situations, there will arise circumstances where proposed 

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connections do not satisfy prequalification requirements. In these situations, the 

requirement for testing reflects the view that the behavior of connections under 

severe cyclic loading cannot be reliably predicted by analytical means alone. 

This suggests that for nonprequalified connections, both laboratory testing 

and the development of an analytical procedure that predicts the behavior are 

required. Requiring an analytical procedure, based on testing, develops a design 

methodology applicable to the design of connections employing slightly different 

members than actually tested. 

Testing is costly and time consuming, and it is the intent of these 

Recommended Criteria to minimize testing requirements to the extent possible. 

Test conditions should match the conditions in the structure as closely as 

possible. 

Figure 6-17 Bolted Bracket Connection 

6.9.1 Testing Procedure 

The testing program should follow the requirements of Appendix S of the 1997 AISC Seismic 

Provisions with the exceptions and modifications discussed below. The program should include 

tests of at least two specimens for a given combination of beam and column size. The results of 

the tests should be capable of predicting the median value of the interstory drift angle capacity 

for the performance states described in Table 6-12. The drift angle capacity q shall be defined as 

indicated in Figure 6-18. Acceptance criteria should be as indicated in Section 6.9.2. 

Table 6-12 Interstory Drift Angle Limits for Various Performance Levels 

Performance Level Symbol Drift Angle Capacity 

Peak Strength qIO Taken as that value of q in Figure 6-18 at which peak load resistance 

occurs. 

Strength degradation qSD Taken as that value of q in Figure 6-18 at which either failure of the 

connection occurs or the strength of the connection degrades to less than 

the nominal plastic capacity, whichever is less 

Ultimate qU Taken as that value of q in Figure 6-18 at which connection damage is so 

severe that continued ability to remain stable under gravity loading is 

uncertain. 

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Column span mid−height to mid−height of story 

DCL 

Beam mid−span 

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Chapter 6: Connection Qualification Steel Moment-Frame Buildings Column span mid−height to mid−height of story

Undeformed centerline 

of beam 

LCL 

q 

D 

= CL 

CL L 

Undeformed centerline 

of beam 

LCL 

q 

D 

= CL 

CL L 

DCL 

Beam mid−span 

Figure 6-18 Drift Angle 

The following modifications and clarifications apply to Appendix S of the 1997 AISC 

Seismic Provisions as modified by Supplement No. 1: 

• In lieu of the requirements in Section S5.2, the size of the beam used in the test specimen 

shall be at least the largest depth and heaviest weight used in the structure. Once the beam is 

chosen, the test column shall be selected to represent properly the inelastic action anticipated 

of the column in the real structure, given the chosen beam. Extrapolation beyond the limits 

stated in this section is not recommended. 

• As an alternative to the loading sequence specified in Section S6.3, the FEMA/SAC loading 

protocol (Krawinkler et al., 2000) is considered acceptable. In the basic loading history, the 

cycles shall be symmetric in peak deformations. The history is divided into steps and the peak 

deformation of each step j is given as qj, a predetermined value of the drift angle. The loading 

history, shown in Table 6-13, is defined by the following parameters: 

qj = the peak deformation in load step j 

nj = the number of cycles to be performed in load step j 

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q

q

1

2

3

4

5

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Table 6-13 Numerical values of qj and nj 

Load Step # Peak deformation qj Number of cycles, nj 

0.00375 6 

0.005 6 

0.0075 6 

0.01 4 

0.015 2 

0.02 2 

0.03 2 

Continue incrementing q in steps of 0.01 radians, and perform two cycles 

at each step until assembly failure occurs. Failure shall be deemed to 

occur when the peak loading falls to 20% of that obtained at qIO or if the 

assembly has degraded to a state at which stability under gravity load 

becomes uncertain. 

Commentary: The AISC Seismic Provisions (AISC, 1997) have been adopted by 

reference into FEMA-302, 1997 NEHRP Recommended Provisions for New 

Buildings. The AISC Seismic Provisions include, and require the use of, 

Appendix S, Qualifying Cyclic Tests of Beam-to-Column and Link-to-Column 

Connections, for qualification of connections that are not pre-qualified. Appendix 

S includes a complete commentary on the requirements. 

Under Appendix S the test specimen must represent the largest beam 

anticipated in the project. The column must be selected to provide a flexural 

strength consistent with the strong-column-weak-beam requirements and panelzone strength requirements. The permitted weight and size limits contained in 

Section S5.2 of Appendix S have been eliminated. 

AISC loading history and acceptance criteria are described in terms of plastic 

rotation while the FEMA/SAC loading protocol, acceptance criteria and design 

recommendations contained in these Recommended Criteria are controlled by 

total drift angle, as previously defined. The engineer should ensure that the 

appropriate adjustments are made when using the AISC loading history with 

these Recommended Criteria. In general, total drift angle is approximately equal 

to plastic rotation plus 0.01 radians. However, the engineer is cautioned that 

plastic rotation demand is often measured in different ways and may require 

transformation to be consistent with the measurements indicated in Figure 6-18. 

The calculation of q illustrated in Figure 6-18 assumes that the top and the 

bottom of the test column are restrained against lateral translation. The height of 

the test specimen column should be similar to that of the actual story height to 

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q

q

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prevent development of unrealistically large contributions to q from flexure of the 

column. 

6.9.2 Acceptance Criteria 

For Simplified Upgrade, the median value of the drift angle capacity at strength degradation, 

qSD, and at connection failure, qU, obtained from qualification testing shall not be less than 

indicated in Table 6-14. The coefficient of variation for these two parameters shall not exceed 

10% unless the mean value, less one standard deviation, is also not less than the value indicated 

in Table 6-14. 

Table 6-14 Minimum Qualifying Total Interstory Drift Angle Capacities, qSD and qU, for 

OMF and SMF Systems 

Structural System Qualifying Drift Angle 

Capacity – Strength 

Degradation, qSD 

(radians) 

Qualifying Drift Angle 

Capacity – Ultimate, qU 

(radians) 

OMF 0.02 0.03 

SMF 0.04 0.06 

Where the clear-span-to-depth ratio of beams in the moment-resisting frame is less than 8, 

the qualifying total drift angle capacities indicated in Table 6-14 shall be increased to q'SD and 

q'U, given by Equations 6-12 and 6-13: 

8d � L L - ¢� qS

¢

D = 

L Ł

�1+ 

L ł

�qSD 

(6-12) 

8d � L L - ¢� qU

¢ = 

L Ł

�1+ 

L �

ł

qU 

(6-13) 

where: q'SD = Qualifying strength degradation drift angle capacity for spans with 

L / d < 8 

qSD = the basic qualifying strength degradation drift angle capacity, in 

accordance with Table 6-14 

q'U = the qualifying ultimate drift angle capacity, for spans with L / d < 8 

qU = the basic qualifying ultimate drift angle capacity, in accordance with Table 

6-14 

L = the center-to-center spacing of columns, per Figure 6-4, inches. 

L' = the distance between points of plastic hinging in the beam, inches. 

d = depth of beam in inches 

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For Systematic Upgrade, the median drift angle capacity for Immediate Occupancy 

performance level shall be taken as the median value of the drift angle, qIO, at which the peak 

connection strength occurs, in accordance with Table 6-12. The median drift angle capacity for 

the Collapse Prevention performance level shall be taken as the median value of the drift angle, 

qU, in accordance with Table 6-12. Resistance factors, f, shall be determined in accordance with 

the procedures of Appendix A of these Recommended Criteria. For any connection, the value of 

f need not be taken as less than 0.75 for the Immediate Occupancy Level or less than 0.5 for the 

Collapse Prevention Level. 

Commentary: This section sets criteria for use in project-specific qualification of 

connection and connection upgrade details, in accordance with Section 6.9 and 

for development of new connection and connection upgrade prequalifications in 

accordance with Section 6.10 of these Recommended Criteria. Two interstory 

drift angle capacities are addressed. The values indicated in Table 6-14 formed 

the basis for extensive probabilistic evaluations of the performance capability of 

various structural systems, reported in FEMA-355F, State of the Art Report on 

Performance Prediction and Evaluation. These probabilistic evaluations indicate 

a high confidence, on the order of 90%, that regular, well-configured frames 

meeting the requirements of FEMA-302 and constructed with connections having 

these capabilities, can meet the intended performance objectives with regard to 

protection against global collapse. They indicate moderate confidence, on the 

order of 50%, that connections can resist Maximum Considered Earthquake 

demands without local life-threatening damage. 

Connection details with capacities lower than those indicated in this section 

may be suitable for upgrades to performance criteria other than those that form 

the basis for FEMA-302. This suitability requires demonstration using the 

performance evaluation procedures contained in Chapter 3 and Appendix A of 

these Recommended Criteria. 

Connections in frames where beam-span-to-depth ratios are less than those 

used for the prequalification testing will experience larger flange strains at the 

plastic hinges, at a particular frame drift, than those tested. For this reason, 

connections used in such frames need to be qualified for larger drifts as indicated 

by Equations 6-12 and 6-13, unless the frames are designed to experience 

proportionally lower drifts than permitted by FEMA-302. 

6.9.3 Analytical Prediction of Behavior 

Connection qualification should include development of an analytical procedure to predict 

the limit states of the connection assembly, as demonstrated by the qualification tests. The 

analytical procedure should permit identification of the strength demands, deformation demands, 

and limit states on various elements of the assembly at the various stages of behavior. The 

analytical procedure should be sufficiently detailed to permit design of connections employing 

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members similar to those tested within the limits identified in Section S5.2 of the 1997 AISC 

Seismic Provisions. 

Commentary: It is important for the designer to have an understanding of the 

limiting behavior of any connection detail so that it may be designed and 

specified on a rational basis for assemblies that vary within specified limits from 

those tested. 

6.10 Prequalification Testing Criteria 

This section provides criteria for development of new prequalifications for connection and 

connection upgrade details for which there is no current prequalification or to extend the 

parametric limitations for prequalification listed in Section 6.5, for general application. 

Prequalification includes a program of connection assembly prototype testing supplemented by a 

suitable analytical procedure that permits prediction of behavior identified in the testing program. 

Commentary: The purpose of this section is to provide recommended procedures 

for prequalification of a connection or connection upgrade detail that is not 

currently prequalified in these Recommended Criteria or to extend the range of 

member sizes that may be used with currently pre-qualified connections for 

general application. These criteria are intended to require significantly more 

testing than are required for a project-specific qualification program, as once a 

connection is prequalified, it can have wide application. Prequalification of a 

connection should incorporate both the testing described in this section and due 

consideration of the following four criteria: 

1. There should be sufficient experimental and analytical data on the 

connection’s performance to establish the likely yield mechanisms and failure 

modes for the connection. 

2. Rational models should be developed and validated for predicting the 

resistance associated with each mechanism and failure mode. 

3. Given the material properties and geometry of the connection, a rational 

procedure should be available to estimate which mode and mechanism 

controls the behavior and the deformation capacity (i.e., the drift angle) that 

can be attained from the controlling conditions. 

4. Given the models and procedures, there should be an adequate data base of 

experiments to permit assessment of the statistical reliability of the 

connection. 

The potential for limit states leading to local collapse (i.e., loss of gravityload capacity) is an important consideration in evaluating the performance of a 

prototype connection. Establishing this limit state as required by Section 6.9.1 

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will necessitate imposing large deformations on the connection. This will require 

loading setups capable of delivering long strokes while withstanding 

correspondingly large out-of-plane deformations or large torsional deformations. 

Many tests are terminated before the ultimate failure of the connection to protect 

the loading apparatus. These early terminations will limit the range over which a 

connection may be prequalified. 

6.10.1 Prequalification Testing 

Testing and acceptance criteria should follow the recommendations in Section 6.9 except that 

at least five nonidentical test specimens shall be used. The resulting range of member sizes that 

will be prequalified should be limited to the range represented by the tested specimens. 

6.10.2 Extending the Limits on Prequalified Connections 

Once a connection has been prequalified, with its parameters lying within certain ranges, 

extending this limitation for general use requires further testing. Testing and acceptance criteria 

should follow the recommendations in Section 6.9 except that at least two nonidentical test 

specimens shall be tested. The resulting range of member size that will be prequalified should be 

limited to those contained in the database of tests for the connection type. 

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Appendix A: Detailed Procedures for Performance Evaluation 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings 

A. Detailed Procedures for Performance Evaluation 

A.1 Scope 

This appendix provides detailed procedures for evaluating the performance capability of steel 

moment-frame buildings. These detailed procedures are provided as a supplement to the 

simplified performance evaluation procedures in Chapter 3. They may be used to demonstrate 

enhanced levels of confidence with regard to the ability of a particular building to meet desired 

performance objectives, relative to the confidence levels that may be derived using the more 

simplified procedures, and they must be used instead of the procedures of Chapter 3 for irregular 

structures and for structures with connections that have not been prequalified. This appendix 

also provides criteria for performance evaluation for deterministically defined hazards. 

Commentary: Chapter 3 provides procedures for a simplified method of 

performance evaluation, using factored-demand-to-capacity ratios to determine a 

level of confidence with regard to a building’s ability to provide a desired 

performance objective. The tabular values of demand and resistance factors and 

confidence indices contained in Chapter 3 were derived using the procedures 

presented in this appendix, applied to the performance evaluation of a suite of 

regularly configured model buildings. Since this suite of model buildings is not 

completely representative of any individual structure, the use of the tabular values 

inherently entails some uncertainty, and thus reduced levels of confidence, with 

regard to performance prediction. The detailed procedures in this appendix 

permit reduction in these uncertainties, and therefore enhanced confidence, with 

regard to prediction of building performance. These more detailed procedures 

must be used for those irregular building configurations not well represented by 

the model buildings used as the basis for the values contained in Chapter 3. 

A.2 Performance Evaluation Approach 

A.2.1 Performance Objectives and Confidence 

As defined in Section 3.2 of these Recommended Criteria, performance is defined in terms of 

probabilistic performance objectives. A performance objective consists of the specification of a 

performance level and an acceptable low probability that poorer performance could occur within 

a specific period of time, typically taken as 50 years. Alternatively, deterministic performance 

objectives can also be evaluated. Deterministic performance objectives consist of the 

specification of a performance level and a specific earthquake, that is, fault location and 

magnitude, for which this performance is to be attained. 

Two performance levels are defined: the Immediate Occupancy performance level and the 

Collapse Prevention performance level. Detailed descriptions of these performance levels may 

be found in Chapter 3. The evaluation procedures contained in this appendix permit estimation 

of a level of confidence associated with achievement of a performance objective. For example, a 

design may be determined to provide a 95% level of confidence that there is less than a 2% 

probability in 50 years of more severe damage than represented by the Collapse Prevention level. 

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For another example, a design may be determined to provide a 50% level of confidence that the 

structure will provide Immediate Occupancy performance, or a better performance, for a Richter 

magnitude 6 earthquake along a defined fault. 

Commentary: The probability that a building may experience damage more 

severe than that defined for a given performance level is a function of two 

principal factors. The first of these is the structure’s vulnerability, that is, the 

probability that it will experience certain levels of damage given that it 

experiences ground motion of certain intensity. The second of these factors is the 

site hazard, that is, the probability that ground shaking of varying intensities may 

occur in a given time period. The probability that damage exceeding a given 

performance level may occur in a period of time is calculated as the integral over 

a year’s time of the probability that damage will exceed that permitted within a 

performance level. Mathematically, this may be expressed as: 

P(D > PL) = � PD>PL (x)h(x)dx (A-1) 

where: 

P(D>PL) = Probability of damage exceeding a performance level in a period 

of t years 

PD>PL(x) = Probability of damage exceeding a performance level given that 

the ground motion intensity is level x, as a function of x, 

h(x)dx = probability of experiencing a ground motion intensity of level (x) to 

(x + dx) in a period of t years 

Vulnerability may be thought of as the capacity of the structure to resist greater 

damage than that defining a performance level. Structural response parameters 

that may be used to measure capacity include the structure’s ability to undergo 

global building drift, maximum tolerable member forces, and maximum tolerable 

inelastic deformations. Ground accelerations associated with the seismic hazard, 

and the resulting enforced global building drift, member forces and inelastic 

deformations produced by the hazard may be thought of as demands. If both the 

demand that a structure will experience over a period of time and the structure’s 

capacity to resist this demand could be perfectly defined, then performance 

objectives, the probability that damage may exceed a performance level within a 

period of time, could be ascertained with 100% confidence. However, the process 

of predicting the capacity of a structure to resist ground shaking demands as well 

as the process of predicting the severity of demands that will actually be 

experienced entail significant uncertainties. Confidence level is a measure of the 

extent of uncertainty inherent in this process. A level of 100% confidence may be 

described as perfect confidence. In reality, it is never possible to attain such 

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Appendix A: Detailed Procedures for Performance Evaluation 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings 

confidence. Confidence levels on the order of 90 or 95% are considered high, 

while confidence levels less than 50% are considered low. 

Generally, uncertainty can be reduced, and confidence increased, by 

obtaining better knowledge or using better procedures. For example, enhanced 

understanding and reduced uncertainty with regard to the prediction of the effects 

of ground shaking on a structure can be obtained by using a more accurate 

analytical procedure to predict the structure’s response. Enhanced 

understanding of the capacity of a structure to resist ground shaking demands can 

be obtained by obtaining specific laboratory data on the physical properties of the 

materials of construction and on the damageability of individual beam-column 

connection assemblies. 

The simplified performance evaluation procedures of Chapter 3 are based on 

the typical characteristics of standard buildings. Consequently, they incorporate 

significant uncertainty in the performance prediction process. As a result of this 

significant uncertainty, it is anticipated that the actual ability of a structure to 

achieve a given performance objective may be significantly better than would be 

indicated by those simple procedures. The more detailed procedures of this 

appendix may be used to improve the definition of the actual uncertainties 

incorporated in the prediction of performance for a specific structure and thereby 

to obtain better confidence with regard to the prediction of performance for an 

individual structure. 

As an example, using the simplified procedures of Chapter 3, it may be found 

that for a specific structure, there is only a 50% level of confidence that there is 

less than a 10% chance in 50 years of poorer performance than the Collapse 

Prevention level. This rather low level of confidence may be more a function of 

the uncertainty inherent in the simplified procedures than the actual inadequate 

capacity of the building to provide Collapse Prevention performance. In such a 

case, it may be possible to use the procedures contained in this appendix to 

reduce the uncertainty inherent in the performance estimation and find that 

instead, there may be as much as a 95% level of confidence in obtaining such 

performance. 

In both the procedures of this appendix and Chapter 3, the uncertainties 

associated with estimation of the intensity of ground motion have been neglected. 

These uncertainties can be quite high, on the order of those associated with 

structural performance or even higher. Thus, the confidence estimated using 

these procedures is really a confidence with regard to structural performance, 

given the seismicity as portrayed by the USGS hazard maps that accompany 

FEMA-273 and FEMA-302. 

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Appendix A: Detailed Procedures for Performance Evaluation Steel Moment-Frame Buildings 

A.2.2 Basic Procedure 

As indicated in Chapter 3, a demand and resistance factor design (DRFD) format is used to 

associate a level of confidence with the probability that a building will have less than a specified 

probability of exceedance of a desired performance level. The basic approach is to determine a 

confidence parameter, l, which may then be used, with reference to Table A-1, to determine the 

confidence level that exists with regard to performance estimation. The confidence parameter, l, 

is determined from the factored-demand-to-capacity equation: 

gg aD l = 

fC (A-2) 

where: 

C = median estimate of the capacity of the structure. This estimate may be obtained either 

by reference to default values contained in Chapters 3 and 6, or by more rigorous 

direct calculation of capacity using the procedures of this appendix, 

D = calculated demand on the structure, obtained from a structural analysis, 

g = a demand variability factor that accounts for the variability inherent in the prediction 

of demand related to assumptions made in structural modeling and prediction of the 

character of ground shaking, 

ga = an analysis uncertainty factor that accounts for the bias and uncertainty associated 

with the specific analytical procedure used to estimate structural demand as a function 

of ground shaking intensity, 

f = a resistance factor that accounts for the uncertainty and variability inherent in the 

prediction of structural capacity as a function of ground shaking intensity, 

l = a confidence index parameter from which a level of confidence can be obtained by 

reference to Table A-1. 

Several structural response parameters are used to evaluate structural performance. The 

primary parameter used for this purpose is interstory drift. Interstory drift is an excellent 

parameter for judging the ability of a structure to resist P-D instability and collapse. It is also 

closely related to plastic rotation demand, or drift angle demand, on individual beam-column 

connection assemblies, and therefore a good predictor of the performance of beams, columns and 

connections. Other parameters used in these guidelines include column axial compression and 

column axial tension. In order to determine a level of confidence with regard to the probability 

that a building has less than a specified probability of exceeding a performance level over a 

period of time, the following steps are followed: 

1. The performance objective to be evaluated is selected. This requires selection of a 

performance level of interest, for example, Collapse Prevention or Immediate Occupancy, 

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and a desired probability that damage in a period of time will be worse than this performance 

level. Representative performance objectives may include: 

• 2% probability of poorer performance than Collapse Prevention level in 50 years 

• 50% probability of poorer performance than Immediate Occupancy level in 50 years. 

It is also possible to express performance objectives in a deterministic manner, where 

attainment of the performance is conditioned on the occurrence of a specific magnitude 

earthquake on an identified fault. 

2. Characteristic motion for the performance objective is determined. For probabilistic 

performance objectives, an average estimate of the ground shaking intensity at the 

probability of exceedance identified in the performance objective definition (step 1) is 

determined. For example, if the performance objective is a 2% probability of poorer 

performance than the Collapse Prevention level in 50 years, then an average estimate of 

ground shaking demands with a 2% probability of exceedance in 50 years would be 

determined. Ground shaking intensity is characterized by the parameter SaT1, the 5% damped 

spectral response acceleration at the site for the fundamental period of response of the 

structure. FEMA-273 provides procedures for determining this parameter for any probability 

of exceedance in a 50-year period. 

For deterministic performance objectives, an average estimate of the ground motion at the 

building site for the specific earthquake magnitude and fault location must be made. As with 

probabilistic estimates, the motion is characterized by SaT1. 

3. Structural demands for the characteristic earthquake ground motion are determined. 

A mathematical structural model is developed to represent the building structure. This model 

is then subjected to a structural analysis, using any of the methods contained in Chapter 3. 

This analysis provides estimates of maximum interstory drift demand, maximum column 

compressive demand, and maximum column-splice tensile demand, for the ground motion 

determined in step 2. 

4. Median estimates of structural capacity are determined. Median estimates of the 

interstory drift capacity of the moment-resisting connections and the building frame as a 

whole are determined, as are median estimates of column compressive capacity and columnsplice tensile capacity. Interstory drift capacity for the building frame, as a whole, may be 

estimated using the default values of Chapter 3 for regular structures, or alternatively, the 

detailed procedures of Section A.6 may be used. These detailed procedures are mandatory 

for irregular structures. Interstory drift capacity for moment-resisting connections that are 

prequalified in Chapters 3 and 6 of these Recommended Criteria may be estimated using the 

default values of Chapters 3 and 6, or alternatively, direct laboratory data on beam-column 

connection assembly performance capability and the procedures of Section A.5 of this 

appendix may be used. Median estimates of column compressive capacity and column-splice 

tensile capacity are made using the procedures of Chapter 3. 

5. A factored-demand-to-capacity ratio, l is determined. For each of the performance 

parameters, i.e., interstory drift as related to global building frame performance, interstory 

drift as related to connection performance, column compression, and column splice tension, 

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Equation A-2 is independently applied to determine the value of the confidence parameter l. 

In each case, the calculated estimates of demand D and capacity C are determined using steps 

3 and 4, respectively. If the procedures of Chapter 3 are used to determine either demand or 

median capacity estimates, than the corresponding values of the demand factors g and 

resistance factors f should also be determined in accordance with the procedures of Chapter 

3. If the procedures of this appendix are used to determine median demand, or capacity, then 

the corresponding demand and resistance factors should be determined in accordance with 

the applicable procedures of this appendix. 

6. Evaluate confidence. The confidence obtained with regard to the ability of the structure to 

meet the performance objective should be the lowest value determined using the values of l 

determined in accordance with step 5 above, back-calculated from the equation: 

l = e 

-bbUT ( KX -k bUT 2) (A-3) 

where: 

b = a coefficient relating the incremental change in demand (drift, force, or 

deformation) to an incremental change in ground shaking intensity, at the 

hazard level of interest, typically taken as having a value of 1.0, 

bUT = an uncertainty measure equal to the vector sum of the logarithmic standard 

deviation of the variations in demand and capacity resulting from uncertainty, 

k = the slope of the hazard curve, in ln-ln coordinates, at the hazard level of 

interest, i.e., the ratio of incremental change in SaT1 to incremental change in 

annual probability of exceedance (refer to Section A.3.2), 

KX = standard Gaussian variate associated with probability x of not being exceeded 

as a function of number of standard deviations above or below the mean found 

in standard probability tables. 

Table A-1 provides a solution for this equation, for various values of the parameters, 

k, l, and bUT. 

The values of the parameter bUT used in Equation A-3 and Table A-1 are used to account for 

the uncertainties inherent in the estimation of demands and capacities. Uncertainty enters the 

process through a variety of assumptions that are made in the performance evaluation process, 

including, for example, assumed values of damping, structural period, properties used in 

structural modeling, and strengths of materials. Assuming that the amount of uncertainty 

introduced by each of the assumptions can be characterized, the parameter bUT can be calculated 

using the equation: 

bUT = 2

ui �i 

b (A-4) 

where: bui are the standard deviations of the natural logarithms of the variation in demand or 

capacity resulting from each of these various sources of uncertainty. Sections A.4, A.5 and A.6 

indicate how to determine bui values associated with demand estimation, beam-column 

connection assembly behavior, and building global stability capacity prediction, respectively. 

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A.3 Determination of Hazard Parameters 

Two basic hazard parameters are required by these performance evaluation procedures. The 

first of these, SaT1, is the median, 5%-damped, linear spectral response acceleration, at the 

fundamental period of the building, at the desired hazard level (probability of exceedance in a 

50-year period or specific earthquake magnitude and fault). Section A.3.1 provides guidelines 

for obtaining this parameter. The second parameter is the slope k of the hazard curve in 

logarithmic space, also evaluated at the desired hazard level. Section A.3.2 provides guidelines 

for obtaining this parameter. 

A.3.1 Spectral Response Acceleration 

Probabilistic, 5%-damped, linear spectral response acceleration, SaT1 at the fundamental 

period of the building, at the desired hazard level (probability of exceedance in a 50-year period), 

may be determined in several different ways. These include: 

a. Site-specific seismological and geotechnical investigation. FEMA-273 provides guidelines 

for this method. 

b. Use of national hazard maps developed by the United States Geologic Survey. FEMA-273 

also provides guidelines for the use of these maps for this purpose. 

Deterministic 5%-damped, linear spectral response acceleration SaT1 at the fundamental 

period of the building, shall be determined based on site-specific seismological and geologic 

study. 

The spectral response acceleration SaT1 is used as a reference point, through which a response 

spectrum is plotted. This response spectrum may be used directly in the structural analysis, or 

alternatively, may be used as a basis for the development of ground motion accelerograms used 

in the structural analysis. Refer to Chapter 3 for guidelines on analysis. 

A.3.2 Logarithmic Hazard Curve Slope 

In these procedures, the logarithmic slope k of the hazard curve at the desired hazard level is 

used to determine the resistance factors, demand factors and also the confidence levels. The 

hazard curve is a plot of probability of exceedance of a spectral amplitude versus that spectral 

amplitude, for a given period, and is usually plotted on a log-log scale. In functional form it can 

be represented by the equation: 

H Si (Si ) = k0 Si 

-k (A-5) 

where: 

HSi(Si) = the probability of ground shaking having a spectral response acceleration 

greater than Si, 

k0 = a constant, dependent on the seismicity of the individual site, 

k = the logarithmic slope of the hazard curve. 

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The slope of the hazard curve is a function of the hazard level, location and response period. 

USGS maps provide values of 5%-damped, spectral response accelerations at periods of 0.2 

seconds, termed Ss, and 1 second, termed S1, for ground motions having 2% and 10% 

probabilities of exceedance in 50 years, for all locations in the U.S. This information is also 

available on their web site and on a CD-ROM. Since most steel moment-frames have relatively 

long fundamental periods, the slope of the hazard curve may be determined for most such 

structures using the S1 values published by the USGS for probabilities of exceedance of 2% and 

10% in 50 years, and substitution of these values into the following equation: 

ln �

� HS1(10/50) �� 

k = Ł

� HS1(2/50) ł� 

= 

1.65 (A-6) 

ln 

�

� S1(2/50) �� ln �� S1(2/50) �� �

Ł S1(10/50) ł

� Ł� S1(10/50) ł� 

where: 

S1(10/50) = spectral amplitude for 10/50 hazard level 

S1(2/50) = spectral amplitude for 2/50 hazard level 

HS1(10/50) = probability of exceedance for 10% in 50 years = 1/475 = 0.0021 

HS1(2/50) = probability of exceedance for 2% in 50 years = 1/2475 = 0.00040 

The accompanying sidebar provides an example of how k may be determined using this 

procedure, for a representative site. As an alternative to using this detailed procedure, an 

approximate value of k may be obtained from Table A-2. When deterministic ground shaking 

demands (specific magnitude earthquake on a fault) are used as the basis for a performance 

objective, the value of k shall be taken as 4.0, regardless of the site seismicity. 

Table A-2 Default Values of the Logarithmic Hazard Curve Slope k 

for Probabilisitc Ground Shaking Hazards 

Region k 

Alaska, California and the Pacific 

Northwest 

3 

Intermountain Region, Basin & 

Range Tectonic Province 

2 

Other U.S. locations 1 

Note: For deterministic ground shaking demands, use a value of k = 4.0 

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Example determination of the parameter, k, the logarithmic slope of the hazard curve using 

hazard data from the USGS. 

Example site location: 

Referencing USGS maps, web site, find S1(10/50)

 = 0.45g, S1(2/50) = 0.77g 

Substituting into equation A-5, find: 

3 .07

0 .537

1 .65

0 .45

0.77 ln 

1 .65

= =

� 

ł 

� � 

Ł 

� = 

g 

g 

k 

Los Angeles City Hall 

A.4 Determination of Demand Factors 

The demand variability factor g and analysis uncertainty factor ga are used to adjust the 

calculated interstory drift, column axial load and column-splice tension demands to their mean 

values, considering the variability and uncertainty inherent in drift demand prediction. 

Variability in drift demand prediction is primarily a result of the fact that due to relatively 

subtle differences in acceleration records, a structure will respond somewhat differently to 

different ground motion records, even if they are well characterized by the same response 

spectrum. Since it is not possible to predict the exact acceleration record that a structure may 

experience, it is necessary to account for the probable variation in demand produced by all 

possible different records. This is accomplished by developing a nonlinear mathematical model 

of the structure, and running nonlinear response history analyses of the structure for a suite of 

ground motion records, all of which are scaled to match the 5% damped linear spectral response 

acceleration, SaT1, described in Section A.3.1. From these analyses, statistics are developed for 

the median value and standard deviation of the natural logarithm of the various demand 

parameters including maximum interstory drift, column axial load, and column splice tension. 

These standard deviations of the natural logarithms of these response parameters are denoted 

bDR. 

Once the value of bDR has been determined, the demand variability factor, g, is calculated 

from the equation: 

k 2 

g = e 2b 

bDR (A-7) 

where: 

k is the logarithmic slope of the hazard curve, taken in accordance with 

Section A.3.1 

b is a coefficient that represents the amount that demand increases as a 

function of hazard, and may normally be taken as having a value of 1.0 

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Uncertainty in the prediction of demands is due to an inability to define accurately the value 

of such parameters as the yield strength of the material, the viscous damping of the structure, the 

effect of nonstructural components, the effect of foundation flexibility on overall structural 

response, and similar modeling issues. Although it is not feasibly practical to do so, it is 

theoretically possible to measure each of these quantities for a building and to model their effects 

exactly. Since it is not practical to do this, instead we use likely values for each of these effects 

in the model, and account for the possible inaccuracies introduced by using these likely values, 

rather than real values. These inaccuracies are accounted for by developing a series of models to 

represent the structure, accounting for the likely distribution of these various parameters. Each 

of these models is used to run analyses with a single ground motion record, and statistics are 

developed for the effect of variation in these parameters on predicted demands. As with the 

variability due to ground motion, the standard deviation of the natural logarithms of the response 

parameters are calculated, and denoted by bDU. This parameter is used to calculate the analytical 

uncertainty factor, ga. 

In addition to uncertainty in demand prediction, the analytical uncertainty factor ga also 

accounts for inherent bias, that is, systematic under- or over-prediction of demand, inherent in an 

analytical methodology. Bias is determined by using the analytical methodology, for example, 

elastic modal analysis, to predict demand for a suite of ground motions and then evaluating the 

ratio of the demand predicted by nonlinear time history analysis of the structure to that predicted 

by the methodology for the same ground motion. This may be represented mathematically as: 

demand predicted by nonlinear time history analysis C (A-8) B = 

demand predicted by analysis method 

where CB is the bias factor. The bias factor that is applicable to a specific structure is taken as 

the median value of CB calculated from a suite of ground motions. The variation in the bias 

factors obtained from this suite of ground motions is used as one of the components in the 

calculation of bDU. 

Once the median bias factor, CB and logarithmic standard deviation in demand prediction bDU 

have been determined, the analysis uncertainty factor, ga is calculated from the equation: 

k 2 

g a = C e 2b 

bDU (A-9) B 

The analysis uncertainty factors presented in Chapter 3 were calculated using this approach 

as applied to a suite of typical buildings. In addition to the uncertainties calculated using this 

procedure, it was assumed that even the most sophisticated methods of nonlinear time history 

analysis entail some uncertainty relative to the actual behavior of a real structure. Additional 

uncertainty was associated with other analysis methods to account for effects of structural 

irregularity, which were not adequately represented in the suite of model buildings used in the 

study. The value of the total logarithmic uncertainty bDU used as a basis for the analysis 

uncertainty factors presented in Chapter 3 are summarized in Table A-3. The bias factors CB 

used in Chapter 3 are summarized in Table A-4. It is recommended that these default values for 

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CB and bDU be used for all buildings. If it is desired to calculate building-specific bDU values, it 

is recommended that these values not be taken as less than those indicated in Table A-3 for 

nonlinear dynamic analysis, for the applicable building characteristics. 

Table A-3 Default Logarithmic Uncertainty bDU for Various Analysis Methods 

Analysis Procedure 

Linear Static Linear 

Dynamic 

Nonlinear 

Static 

Nonlinear 

Dynamic 

Performance Level IO CP IO CP IO CP IO CP 

Type 1 Connections 

Low Rise (<4 stories) 0.17 0.22 0.15 0.16 0.14 0.17 0.10 0.15 

Mid Rise (4 – 12 stories) 0.18 0.29 0.15 0.23 0.15 0.23 0.13 0.20 

High Rise (> 12 stories) 0.31 0.25 0.19 0.29 0.17 0.27 0.17 0.25 

Type 2 Connections 

Low Rise (<4 stories) 0.19 0.23 0.16 0.25 0.18 0.18 0.10 0.15 

Mid Rise (4 – 12 stories) 0.20 0.30 0.17 0.33 0.14 0.21 0.13 0.20 

High Rise (> 12 stories) 0.21 0.36 0.21 0.31 0.18 0.33 0.17 0.25 

Table A-4 Default Bias Factors CB 

Analysis Procedure 

Linear Static Linear 

Dynamic 

Nonlinear 

Static 

Nonlinear 

Dynamic 

Performance Level IO CP IO CP IO CP IO CP 

Type 1 Connections 

Low Rise (<4 stories) 0.90 0.65 1.00 0.80 1.10 0.85 1.00 1.00 

Mid Rise (4 – 12 stories) 1.10 0.85 1.10 1.15 1.40 0.95 1.00 1.00 

High Rise (> 12 stories) 1.05 1.0 1.15 1.0 1.30 0.85 1.00 1.00 

Type 2 Connections 

Low Rise (<4 stories) 0.75 0.90 1.00 1.20 0.90 1.25 1.00 1.00 

Mid Rise (4 – 12 stories) 0.80 1.00 1.05 1.30 1.08 1.35 1.00 1.00 

High Rise (> 12 stories) 0.75 0.70 1.30 1.20 1.30 1.30 1.00 1.00 

Commentary: Although it may be possible, for certain structures, to increase the 

confidence associated with a prediction of probable earthquake demands on the 

structure, through calculation of structure-specific analysis uncertainty factors, in 

general this is a very laborious process. It is recommended that the default 

values of bDU and CB, contained in Tables A-3 and A-4, be used for most 

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structures. However, the procedures of this section can be used to adjust the 

analysis uncertainty and demand variability factors for the site seismicity k. 

A.5 Determination of Beam-Column Connection Assembly Capacities 

The probable behavior of beam-column connection assemblies at various demand levels can 

best be determined by full-scale laboratory testing. Such testing can provide indications of the 

probable physical behavior of such assemblies in buildings. Depending on the characteristics of 

the assembly being tested, meaningful behaviors may include the following: onset of local 

buckling of flanges; initiation of fractures in welds, base metal or bolts; a drop in the moment 

developed by the connection beyond predetermined levels; or complete failure, at which point 

the connection is no longer able to maintain attachment between the beam and column under the 

influence of gravity loads. If sufficient laboratory data are available, it should be possible to 

obtain statistics, including a median value and standard deviation, on the demand levels at which 

these various behaviors occur. 

In the past, most laboratories used plastic rotation as the demand parameter by which beamcolumn connection assembly behavior was judged. However, since plastic deformation may 

occur at a number of locations within a connection assembly, including within the beam itself, 

within the connection elements, and within the column panel zone or column, many laboratories 

have measured and reported plastic rotation angles from testing in an inconsistent manner. 

Therefore, in these Recommended Criteria, total interstory drift angle, as indicated in Section 

3.6, is the preferred demand parameter for reporting laboratory data. This parameter is less 

subject to erroneous interpretation by testing laboratories and also has the advantage that it is a 

quantity directly predicted by linear structural analyses. 

Median drift angle capacities, C, and resistance factors, f, for various prequalified 

connection types are presented in Chapters 3 and 6. These values were determined from cyclic 

tests of full-size connection assemblies using the testing protocols indicated in Section 6.9. The 

cyclic tests are used to determine the load-deformation hysteresis behavior of the system and the 

connection drift angle at which the following behaviors occur: 

1. onset of local flange buckling of beams, 

2. degradation of moment-resisting capacity of the assembly to a value below the nominal 

moment-resisting capacity, 

3. initiation of fracture of bolts, welds, or base metal that results in significant strength 

degradation of the assembly, and 

4. complete failure of the connection, characterized by an inability of the connection to 

maintain its integrity under gravity loading. 

Based on this data, drift angle statistics, including a median value and logarithmic standard 

deviation are obtained for the Immediate Occupancy and Collapse Prevention damage states, as 

indicated in Table A-5. The quantity qU, the ultimate capacity of the connection, is used to 

evaluate the acceptability of connection behavior for the Collapse Prevention performance level 

as limited by local behavior. 

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Table A-5 Behavior States for Performance Evaluation of Connection Assemblies 

Symbol Performance Level Description 

qIO Immediate Occupancy The lowest drift angle at which any of behaviors 1, 2, or 3, occur 

(see Section A.5, above) 

qU Ultimate The drift angle at which behavior 4 occurs 

qSD Strength Degradation The lowest drift angle at which any of behaviors 2, 3, or 4 occur 

A.5.1 Connection Test Protocols 

Two connection test protocols have been developed under this project. The standard protocol 

is intended to represent the energy input and cyclic deformation characteristics experienced by 

connection assemblies in steel moment frames which are subjected to strong ground shaking 

from large magnitude earthquakes, but which are not located within a few kilometers of the fault 

rupture. This protocol presented in Section 6.9 is similar to that contained in ATC-24 and 

consists of ramped cyclic loading, starting with initial cycles of low energy input within the 

elastic range of behavior of the assembly, and progressing to increasing deformation of the beam 

tip until assembly failure occurs. However, unlike ATC-24, the protocol incorporates fewer 

cycles of large-displacement testing to balance more closely the energy input to the assembly, 

with that likely experienced by framing in a real building. The second protocol is intended to 

represent the demands experienced by connection assemblies in typical steel moment-frame 

buildings responding to near-fault ground motion, dominated by large velocity pulses. This 

protocol (Krawinkler, 2000) consists of an initial single large displacement, representing the 

initial response of a structure to a velocity pulse, followed by repeated cycles of lesser 

displacement. 

Performance characteristics of connection assemblies, for use in performance evaluation of 

buildings, should be selected based on the characteristics of earthquakes dominating the hazard 

for the building site, at the specific hazard level. Most buildings are not located on sites that are 

likely to be subjected to ground shaking with near-field pulse characteristics. Connection 

performance data for such buildings should be based on the standard protocols. Buildings on 

sites that are close to a major active fault are most likely to experience ground shaking with these 

strong pulse-like characteristics and connection performance for such buildings should be based 

on the near-fault protocol. However, qualification of connections for classification as either 

Type 1 or Type 2 connections should be based on the standard protocol. 

A.5.2 Determination of Beam-Column Assembly Capacities and Resistance Factors 

Median drift angle capacities for the quantities qIO and qU should be taken directly from 

available laboratory data. The median value should be taken as that value from all of the 

available tests that is not exceeded by 50% of the tests. The value of the quantity f, for each of 

the Immediate Occupancy and ultimate (Collapse Prevention) states should be determined by the 

following procedure. 

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1. Obtain the logarithmic standard deviation of the qIO or qU values available from the 

laboratory data. That is, take the standard deviation of the natural logarithms of the qIO or qU 

values respectively, obtained from each laboratory test. Logarithmic standard deviation may 

be determined from the formula: 

b = 

( )

1 

ln ln 

1 

2 

-

� - =

n 

x x n 

i i i (A-10) 

where: 

b = the standard deviation of the natural logarithms of the test data

xi = individual test data value

n = the number of tests from which data is available

ln xi =the mean of the logarithms of the xi values. 

2. Calculate the connection resistance factor fR due to randomness, the observed variation in 

connection behavior, from laboratory testing, using the equation: 

k b 2 -

fR = e 2b (A-11) 

where:

k = the slope of the hazard curve, determined in accordance with Section A.3.2

b = a coefficient that relates the change in hazard to the change in demand, and which

may be taken as having a value of 1.0 

b = the logarithmic standard deviation calculated in accordance with Equation A-10. 

3. Determine the connection resistance factor accounting for random and uncertain behaviors 

from the equation: 

k 

f = fRfU = fRe 

-

2b

(0.2)

2 

(A-12) 

where: 

fR = the resistance factor accounting for random behavior 

fU = the resistance factor accounting for uncertainty in the relationship between 

laboratory findings and behavior in real buildings, and assumed in these 

Recommended Criteria to have a logarithmic standard deviation bu of 0.2 

A.6 Global Stability Capacity 

For the Collapse Prevention performance level, in addition to consideration of local behavior, 

that is, the damage sustained by individual beams and beam-column connection assemblies, it is 

also important to consider the global stability of the frame. The procedures indicated in this 

section are recommended for determining an interstory drift capacity C and resistance factor f 

associated with global stability of the structure. 

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The global stability limit is determined using the Incremental Dynamic Analysis (IDA) 

technique. This requires the following steps: 

1. Choose a suite of ten to twenty accelerograms representative of the site and hazard level for 

which the Collapse Prevention level is desired to be achieved. 

2. Select one of these accelerograms and perform an elastic time-history analysis of the 

building. Determine a scaling factor for this accelerogram such that the elastic time history 

analysis would result in response that would produce incipient yielding in the structure. 

Determine the 5%-damped, spectral response acceleration SaT1 for this scaled accelerogram at 

the fundamental period of the structure. On a graph with an abscissa consisting of peak 

interstory drift and an ordinate axis of SaT1, plot the point consisting of the maximum 

calculated interstory drift from the scaled analysis and the scaled value of SaT1. Draw a 

straight line from the origin of the axes to this point. The slope of this line is referred to as 

the elastic slope, Se 

3. Increase the scaling of the accelerogram, such that it will produce mild nonlinear behavior of 

the building. Perform a nonlinear time-history analysis of the building for this scaled 

accelerogram. Determine the SaT1 for this scaled accelerogram and the maximum predicted 

interstory drift from the analysis. Plot this point on the graph. Call this point D1. 

4. Increase the scaling amplitude of the accelerogram slightly and repeat Step 3. Plot this point 

as D2. Draw a straight line between points D1 and D2. 

5. Repeat Step 4 until the straight line slope between consecutive points Di and Di+1, is less than 

0.2 Se. When this condition is reached, Di+1 is the global drift capacity for this accelerogram. 

If Di+1 > 0.10 then the drift capacity is taken as 0.10. Figure A-1 presents a typical series of 

plots obtained from such analyses. 

6. Repeat Steps 2 through 5 for each of the accelerograms in the suite selected as representative 

of the site and hazard and determine an interstory drift capacity for the structure for each 

accelerogram. 

7. Determine a median interstory drift capacity C for global collapse as the median value of the 

calculated set of interstory drift capacities, determined for each of the accelerograms. The 

median value is that value exceeded by 50% of the accelerograms. 

8. Determine a logarithmic standard deviation b for random differences in ground motion 

accelerograms, using Equation A-10 of Section A.5.2. In this equation, xi is the interstory 

drift capacity predicted for the i th accelerogram, and n is the number of accelerograms 

contained in the analyzed suite. 

9. Calculate the global resistance factor fR due to randomness in the predicted global collapse 

capacity for various ground motions from the equation: 

k b 2 -

fR = e 2b (A-13) 

where k and b are the parameters described in Section A.5.2 and b is the logarithmic standard 

deviation calculated in the previous step. 

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0 

0.5 

1 

1.5 

2 

2.5 

0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 

Max im um Interstory Drift 

Spectral Acceleration (at fundame ntal period), g 

LA23 

LA22 

LA30 

LA24 

LA28 

Figure A-1 Representative Incremental Dynamic Analysis Plots 

10. Determine a resistance factor for global collapse from the equation: 

k 2 -

f = fUfR = e 2b 

bU fR (A-14) 

where: 

fR is the global resistance factor due to randomness determined in Step 9. 

bU is the logarithmic standard deviation related to uncertainty in analytical prediction of 

global collapse prevention taken as having a value of 0.15 for low-rise structures, 3 

stories or less in height; a value of 0.2 for mid-rise structures, 4 stories to 12 stories in 

height; and taken as having a value of 0.25 for high-rise structures, greater than 12 stories 

in height. 

It is important that the analytical model used for determining the global drift demand be as 

accurate as possible. The model should include the elements of the moment-resisting frame as 

well as framing that is not intended to participate in lateral load resistance. A nominal viscous 

damping of 3% of critical is recommended for most buildings. The element models for beamcolumn assemblies should realistically account for the effects of panel zone flexibility and 

yielding, element strain hardening, and stiffness and strength degradation, so that the hysteretic 

behavior of the element models closely matches that obtained from laboratory testing of 

comparable assemblies. 

Commentary: As noted above, accurate representation of the hysteretic behavior 

of the beam-column assemblies is important. Earthquake-induced global collapse 

initiates when displacements produced by the response to ground shaking are 

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Appendix A: Detailed Procedures for Performance Evaluation Steel Moment-Frame Buildings 

large enough to allow P-D instabilities to develop. Prediction of the onset of P-D 

instability due to ground shaking is quite complex. It is possible that during an 

acceleration record a structure will displace to a point where static P-D 

instability would initiate, only to have the structure straighten out again before 

collapse can occur, due to a reversal in ground shaking direction. 

The basic effect of P-D instability is that a negative tangent stiffness is 

induced in the structure. That is, P-D effects produce a condition in which 

increased displacement can occur at a reduced lateral force. A similar and 

equally dangerous effect can be produced by local hysteretic strength degradation 

of beam-column assemblies (FEMA-355C). Hysteretic strength degradation 

typically occurs after the onset of significant local buckling in the beam-column 

assemblies. It is important when performing Incremental Dynamic Analyses that 

these local strength degradation effects, which show up as a concave curvature in 

the hysteretic loops in laboratory data, are replicated by the analytical model. 

Nonlinear analysis software that is currently commercially available is not, in 

general, able to model this behavior. These effects can be approximately 

accounted for by increasing the amount of dead load on the structure, to produce 

artificially the appropriate negative stiffness. 

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Appendix B: Detailed Procedures for Loss Estimation 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings 

B. Detailed Procedures for Loss Estimation 

B.1 Introduction 

This appendix describes detailed loss estimation procedures for developing structural damage 

functions and related direct economic loss functions for welded, steel moment-frame (WSMF) 

buildings. These procedures are compatible with the HAZUS (NIBS, 1997a) methodology, a 

complex collection of modules that work together to estimate casualties, loss of function and 

economic impacts on a region due to a scenario earthquake. The HAZUS methodology was 

developed for the Federal Emergency Management Agency (FEMA) by the National Institute of 

Building Sciences (NIBS) and is documented in a three-volume Technical Manual (NIBS, 

1997b). One of the main components of the methodology estimates the probability of various 

states of structural and nonstructural damage to buildings. Other modules of the methodology 

use the damage state probabilities to estimate various types of building-related losses. The 

HAZUS methodology is intended primarily for use in estimation of earthquake losses in regions 

with a large inventory of buildings represented by generic building types. 

The procedures presented in this appendix utilize the results of WSMF building performance 

evaluations conducted in accordance with Chapter 3 of these Recommended Criteria, 

supplemented by default values of parameters provided in this appendix, to construct structural 

damage and loss functions. Specifically, structural analysis using the nonlinear static method 

must be performed as a precursor to the application of the loss estimation methods presented 

herein. Default values of damage and loss parameters are provided for typical 3-story, 9-story 

and 20-story WSMF buildings. Example loss estimates that illustrate application of the detailed 

methods are developed for typical 9-story WSMF buildings. 

Commentary: To support mitigation efforts, FEMA funded NIBS to develop 

“Procedures for Development of HAZUS-Compatible Building-Specific Damage 

and Loss Functions” (Kircher, 1999). These procedures are an extension of the 

more general methods of HAZUS, but allow users to incorporate building-specific 

data including capacity and fragility values developed by nonlinear static 

(pushover) analysis of the building of interest. The purpose of such evaluations is 

to understand better the response behavior of the structure, the modes of 

structural damage and failure, and the amount of structural damage (e.g., 

connection damage) as a function of the level of earthquake ground shaking. 

These so-called “building-specific” methods provide the primary basis for the 

detailed loss-estimation procedures of this appendix. 

Implementation of the detailed procedures requires users to have certain 

levels of expertise and knowledge. It is anticipated that users will be structural 

engineers: 

1. familiar with evaluation of the earthquake behavior of buildings, 

2. experienced with nonlinear building analysis, 

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3. familiar with basic methods of statistical analysis, and 

4. familiar with the HAZUS methodology and building-specific procedures. 

In addition to the HAZUS Technical Manual (NIBS, 1997b), further 

references on the HAZUS methodology may be found in papers contained in a 

1997 special issue of Earthquake Spectra on loss estimation published by 

Earthquake Engineering Research Institute (EERI). Pertinent papers include 

Whitman et al. (1997), and Kircher et al. (1997a,b). 

B.2 Scope 

B.2.1 General 

The scope of the detailed loss-estimation procedures is limited to steel moment-frame 

(WSMF) building damage caused by ground shaking. While ground shaking typically dominates 

earthquake loss, other hazards, such as ground failure, due to either liquefaction or land-sliding, 

and surface fault rupture, can also cause building damage. Although less prevalent, when 

building damage due to ground failure or surface fault rupture occurs it is typically more severe 

than building damage caused by ground shaking. 

The scope of detailed loss-estimation procedures is further limited to damage to the structural 

system of WSMF buildings. While structural (connection-related) damage is the primary focus 

of this report, significant damage and loss can occur to nonstructural components and to building 

contents. Typically, at lower states of damage, nonstructural and contents losses are greater, by 

several times, than structural losses. This is due to the fact that damage usually begins to occur 

in nonstructural systems and can become severe before any damage occurs to the structural 

system. At higher states of damage, the structure becomes more important to economic loss 

estimation since damage to the structure can affect a complete loss of both structural and 

nonstructural systems (and contents), and cause long-term closure of the building (that is, loss of 

function). 

The scope of detailed loss-estimation procedures is still further limited to direct economic 

losses associated with repair and replacement of damaged structural elements and to building loss 

of function. 

Commentary: Other types of losses, such as casualties, may also be important to 

the user. In those cases for which users require loss estimates for hazards other 

than ground shaking, the HAZUS Technical Manual (NIBS, 1997b) should be 

used to develop appropriate loss models. In those cases for which users require 

loss estimates for building damage other than structural and loss types other than 

economic, Kircher (1999) should be used to augment the detailed procedures of 

this section. 

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Appendix B: Detailed Procedures for Loss Estimation 

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Steel Moment-Frame Buildings 

B.2.2 Typical Welded, Steel Moment-Frame (WSMF) Buildings 

Detailed loss-estimation methods permit the development of building-specific loss functions, 

based on the configuration and structural details of a specific building. In order to allow more 

general application, this appendix also presents a series of default loss functions, derived using 

these methods for use in prediction of damage to WSMF buildings of different height, different 

seismic force design and different connection type, without needing to resort to detailed 

structural analyses of individual buildings. Default values of various damage and loss parameters 

are provided for typical 3-story, 9-story and 20-story buildings. Default values are provided for 

buildings located in different regions (having different design codes and practice) and having 

different connection conditions, as identified in Table B-1. 

Table B-1 Connections in Typical WSMF Buildings in Three Regions 

Connection Condition Los Angeles 

Region 

Seattle Region Boston 

Region 

Pre-Northridge X X X 

Post-Northridge 

Special Moment Frame (SMF) 

X X 

Damaged Pre-Northridge X 

A pre-Northridge connection condition assumes that the building has beam-column 

connections typical of buildings designed and built prior to the 1994 Northridge earthquake, but 

which have not been damaged by earthquake ground shaking. A post-Northridge connection 

condition assumes that the building has either new or retrofitted beam-column connections that 

comply with the recommendations of FEMA-350 Recommended Seismic Design Criteria for 

New Steel Moment-Frame Buildings, as applied to Special Moment-Resisting Frame Systems. A 

damaged pre-Northridge connection condition assumes the building has beam-column 

connections that are typical of pre-Northridge buildings and that have sustained substantial 

earthquake damage, but have not been repaired. 

B.3 Damage States 

Structural damage is described by one of four discrete damage states: Slight, Moderate, 

Extensive and Complete. Of course, actual building damage varies as a continuous function of 

earthquake demand. Ranges of damage are used to describe damage, since it is not practical to 

have a continuous scale, and damage states provide users with an understanding of the structure’s 

physical condition. Descriptions of structural damage states for WSMF buildings (HAZUS 

model building type S1), based upon but modified from the HAZUS Technical Manual are 

indicated in Table B-2. 

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Table B-2 Descriptions of Structural Damage States 

Damage State Buildings with Pre-Northridge 

Connections 

Buildings with Post-Northridge 

Connections 

Slight structural damage No permanent interstory drift. 

Minor deformations in some 

connection elements and fractures 

in less than 10% of the 

connections at any floor level. 

No permanent interstory drift. 

Minor deformations in some 

connection elements. No 

fractures in connections. 

Moderate structural damage Permanent interstory drift as large 

as 0.5%. Perhaps as many as 

25% of the connections on any 

floor level have experienced 

fracture. 

Permanent interstory drift as large 

as 0.5%. Moderate amounts of 

yielding and distortion of some 

column panel zones. Minor 

buckling of some girders. 

Extensive structural damage Many connections have failed 

with a number of fractures 

extending into and across column 

panel zones. Some connections 

may have lost ability to support 

gravity load, resulting in partial 

local collapse. Large permanent 

interstory drifts occur in some 

stories. 

Many steel members have 

exceeded their yield capacity, 

resulting in significant permanent 

lateral deformation of the 

structure. Some structural 

members or connections may 

have major permanent member 

rotations at connections, buckled 

flanges and failed connections. 

Some connections may have lost 

ability to support gravity load, 

resulting in partial local collapse. 

Complete structural damage A significant portion of the structural elements have exceeded their 

ultimate capacities and/or many critical structural elements or 

connections have failed resulting in dangerous permanent lateral 

displacement, partial collapse or collapse of the building. 

Approximately 15% (of the total square footage) of all WSMF 

buildings with complete damage are expected to have collapsed. 

General guidance to users regarding selection of damage parameters, taken from Kircher 

(1999), is provided in Table B-3. Additional steel moment-frame (WSMF) building-specific 

guidance is given in Table B-4 for determining the structural damage state based on the fraction 

of damaged connections. 

Table B-3 General Guidance for Expected Loss Ratio and Building Condition in Each 

Damage State 

Damage State 

Likely Amount of Damage, Loss, or Building Condition 

Range of 

Possible Loss 

Ratios 

Probability of 

Long-Term 

Building Closure 

Probability of 

Partial or Full 

Collapse 

Immediate 

Postearthquake 

Inspection 

Slight 0% - 5% P = 0 P = 0 Green Tag 

Moderate 5% - 25% P = 0 P = 0 Green Tag 

Extensive 25% - 100% P @ 0.5 P @ 01 Yellow Tag 

Complete 100% P @ 1.0 P > 0 Red Tag 

1. Extensive damage may include some localized collapse of the structure. 

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Table B-4 Specific Guidance for Selection of Damage State Based on Connection Damage 

Fraction of All Connections Likely to be Damaged1 

Average Fraction Fraction Range Damage State

0.02 0.0 – 0.05 Slight 

0.10 0.05 – 0.25 Moderate 

0.50 0.25 – 0.75 Extensive 

@1.0 0.75 – 1.0 Complete 

1. Connections having indications of flaws at the root of the Complete Joint Penetration (CJP) 

weld of beam flanges to columns are not considered as having damage. 

B.4 Basic Approach 

For the detailed procedures, maximum interstory drift is the basic parameter used to assess 

structural (i.e., connection) damage. Based on the calculated maximum interstory drift demand, 

the probability that a structure will be damaged sufficiently to be classified as conforming to each 

of the four damage states described in Section B.3, is determined. For example, at a maximum 

interstory drift demand of 3%, a structure may be found to have a low probability, only 10%, of 

having only slight damage, a 30% probability of moderate damage, a 40% probability of 

extensive damage and a 20% probability of complete damage. This probabilistic approach is 

taken in recognition of the fact that due to inherent uncertainties in the prediction of ground 

motion, structural response and structural damage, it is not possible to quantify precisely how 

much damage a structure will have for a given earthquake. In this methodology, the probabilistic 

relationship between structural damage and maximum interstory drift is termed a fragility 

function. Fragility functions are defined by median estimates of the maximum interstory drift at 

which a damage state will initiate in a structure (damage state medians) and a parameter b that 

represents the uncertainty associated with these estimates. 

Maximum interstory drift is defined as the peak drift (throughout the duration of earthquake 

shaking) occurring in any story in the building. Maximum interstory drift is assumed to be about 

the same as the drift angle demand on nearby beam-column connections. On this basis, damage 

states of buildings with pre-Northridge connection conditions are related (and calibrated) to 

observed building response and damage. Similarly, users can define damage states (fragility 

medians) of buildings with post-Northridge connection conditions using the results of laboratory 

testing of connections. 

In general, the maximum interstory drift in a structure will be greater than the average drift 

calculated over the height of the building due to various building characteristics (e.g., modes of 

vibration, nonlinearity, etc.) and the specific nature of the earthquake ground shaking. While 

response history analyses (of complex multi-degree-of-freedom nonlinear models) provide the 

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most accurate and complete set of building response data, such analyses are rarely practical for 

engineering applications and are not required for this methodology. 

The detailed procedures rely on nonlinear static (pushover) analysis to estimate peak 

interstory drift and damage. Height-dependent factors are used to adjust pushover drift results 

for higher-mode effects and other effects not explicitly included in the nonlinear static analysis. 

Similarly, other height-dependent modal factors are used to relate maximum interstory drift to 

spectral displacement demand, so that damage (fragility) functions may be expressed in terms of 

spectral displacement but still be based on the drift angle limits of connections at the story (or 

stories) experiencing the maximum drift. 

The overall approach or process used to estimate economic loss involves a number of steps, 

as illustrated in the flowchart of Figure B-1. Users are expected to select an appropriate scenario 

earthquake and to develop the 5%-damped response spectrum of this earthquake using, for 

example, the generalized spectrum shape and soil amplification factors described in FEMA-273 

or FEMA-302. 

Peak Response 

Determine peak spectral 

displacement – intersection of 

capacity and demand 

Shaking Demand 

Define scenario earthquake and 

develop 5%-damped response 

spectrum 

Structure Capacity 

Develop capacity curve of 

structure using pushover 

analysis results 

Damage Probability 

Determine damage state 

probabilities – intersection of peak 

response and structure fragility 

Mean Earthquake Losses 

Estimate economic and functional 

losses – combine probability of 

damage and loss functions 

Structure Fragility 

Develop fragility curves of 

structure using pushover 

analysis results 

Loss Functions 

Develop economic loss 

functions using building data 

and other information 

Figure B-1 Flowchart of Detailed Loss Estimation 

Users are also expected to provide other information and data for the building. This can 

range from basic structural data obtained from the construction documents to results obtained 

from a nonlinear static analysis of the building, conducted in accordance with the guidelines of 

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Steel Moment-Frame Buildings 

Chapter 3. Section B.5 summarizes required input data to be supplied by the user. Subsequent 

sections provide guidance for developing structure capacity, structure response, structure fragility 

and building loss functions. 

B.5 Required Data — User Input 

The accuracy of loss estimates performed using the detailed methodology depends primarily 

on the extent and quality of the information provided by the user. While default data is provided 

and may be used if considered appropriate, the more effort the user puts into the determination of 

building data, the more reliable the results will be. 

It is expected that the user will have seismic hazard data available. Although not required for 

development of damage and loss functions, seismic hazard data, including site soil conditions, 

are important and must be input by the user when developing loss estimates. It is also expected 

that, as a minimum, the user will have basic data on the building characteristics, such as the 

building size, occupancy (that is, use, rather than the number of occupants) and replacement cost. 

Users are expected to calculate a pushover curve for the building at displacements up to 

complete failure of the structure. This may require pushing the building beyond the target 

displacement used in performance evaluation, as in Chapter 3, particularly if the evaluated 

performance objective was based on a low hazard level. The pattern of applied lateral loading 

should be based on the fundamental mode in the direction of interest and pushover results should 

represent both horizontal directions of building response (i.e., both principal axes of the 

building). If pushover results are significantly different for the two different directions, separate 

pushover curves should be developed and used to estimate losses for each direction. Threedimensional models that permit rotation as well as translation should be used for pushover 

analysis of structures with plan irregularities that affect torsion. 

Users are expected to have an understanding of the expected performance of the components 

of the structural system and the modes of failure as a function of building interstory drift. In 

addition to drift, Chapter 3 has identified other key performance parameters including column 

axial-load capacity and column tension-splice capacity that should be considered when 

determining at what drift level various failure modes and damage states are expected to occur. 

Users are expected to provide the total replacement value of the structural system, expressed 

in terms of dollars/square foot. Although not required (default values are included in this 

appendix), users should also provide input on the repair of structural damage. That is, for each 

damage state, the user could review the associated damage to the structure and develop a cost and 

schedule for elements and components requiring repair. This may be done judgmentally, or more 

thoroughly by developing actual repair schemes, and obtaining estimates of, for example, 

construction costs, schedule, and building interruption. 

B.5.1 Building Capacity Curve 

The building capacity curve is derived from the pushover curve using modal properties for 

the building and a standard shape compatible with the HAZUS methodology. Specifically, the 

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capacity curve is the pushover curve transformed from coordinates of base shear and roof 

displacement to coordinates of spectral acceleration (SA) and spectral displacement (SD). This 

coordinate transformation is accomplished on a point by point basis, by using the formulas: 

SDi = a 2 Di (B-1) 

V S Ai = 

i W (B-2) 

a1 

where: a1 = fraction of building weight effective in the fundamental mode 

in the direction under consideration (Equation B-3), 

a2 = fraction of building height at the elevation where the fundamental-modal 

displacement is equal to spectral displacement (Equation B-4), 

Di = displacement at point “i” on the pushover curve, 

Vi = base shear force at point “i” on the pushover curve (kips), 

W = building weight (kips), 

and: 

2

� N 

��(wi

f ip ) g �

� 

Ł i=1 ł a1 = Ø N øØ N ø (B-3) 

Œ�(wi ) g œŒ�(wi

f 2 

ip ) g œ o i=1 ßo i=1 ß 

N 

�(wi

f 2 

ip ) g 1 i=1 a 2 = = (B-4) PF f Ø N ø p cp,p Œ�(wi

fip ) g œfcp,p o i=1 ß 

where: wi / g = mass assigned to the i

th degree of freedom, 

fip = amplitude of modal shape at i

th degree of freedom, 

fcp,p = amplitude of mode shape at control point, 

N = number of degrees of freedom. 

Some structural analysis software programs have the capability of automatically converting 

pushover curves to capacity curves using these formulas. As a simpler approximation to the 

formulas for a1 and a2 given above, these modal factors may be reasonably well estimated based 

only on the number of stories, N, using the following formula: 

1 

@ 

1 

@ N 0.14 £ 1.5 (B-5) 

a1 a 2 

In the HAZUS methodology, two control points define a standard shape for the capacity 

curve. These are the yield capacity control point and the ultimate capacity control point, as 

shown in Figure B-2. The yield point (normally designated by Dy, Ay) defines the limit of the 

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elastic domain and the ultimate point (normally designated by Du, Au) defines the point along the 

curve where the structure is assumed to be fully plastic. 

The user is expected to define capacity curve control points from the actual capacity curve 

using both judgment and the following rules: 

• Yield capacity control point (Dy, Ay) is selected as the point where significant yielding is just 

beginning to occur (slope of capacity curve is essentially constant up to the yield point). 

• The expected period, Te, of the building, at or just below yield, should be the true “elastic” 

fundamental-mode period of the building: 

Te @ 0.32 

y 

y 

A 

D 

(B-6) 

• The ultimate capacity control-point acceleration, Au, is selected as the point of maximum 

spectral acceleration (maximum building strength), not to exceed the value of spectral 

acceleration at which the structure has just reached its full plastic capacity. 

• The ultimate capacity control-point displacement, Du, is selected as the greater of either the 

spectral displacement at the point of maximum spectral acceleration or the spectral 

displacement corresponding to Equation B-7: 

Au Du = 2Dy A (B-7) 

y 

Yield Capacity 

Control Point 

Ultimate Capacity Control Point 

(at fully plastic state) 

HAZUS-Compatible 

Capacity Curve (dashed) 

Capacity Curve 

(Spectral Accel. vs. Spectral Disp.) 

Normalized Pushover Curve 

(Base Shear/W vs. Roof Disp.) 

(V/W, DCP) 

(SD, SA) 

SD = DCP a2 

SA = (V/W) /a1 

Displacement 

Figure B-2 Example Development of Standard (HAZUS-Compatible) Capacity Curve 

from a Normalized Pushover Curve 

Commentary: The HAZUS definition of the elastic period Te is the same as the 

initial period, and must not be confused with the definition of the effective period 

Te contained in FEMA-273. The effective period Te of FEMA-273 is based on Acceleration 

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Appendix B: Detailed Procedures for Loss Estimation Steel Moment-Frame Buildings 

stiffness at 60% of the ultimate strength of the building and should not be used for 

loss estimation since it generally overestimates the displacement of the building. 

Table B-5 summarizes the elastic period and capacity curve control points for 

typical steel moment-frame buildings studied in this project. Capacity was 

derived from pushover analyses using modal properties based on Equation B-5. 

Building period and pushover properties were based on analyses reported in 

FEMA-355C and pertain to buildings conforming to the 1994 Uniform Building 

Code requirements. Individual buildings conforming to these same code 

provisions may be either stronger or weaker than those analyzed and buildings 

designed to other code requirements are likely to have substantially different 

characteristics than those indicated. 

Table B-5 Capacity Curve Properties of Typical Welded Steel Moment-Frame Buildings 

Capacity 

Parameter 

Pre-Northridge Connections Post-Northridge Connections 

3-Story 9-Story 20-Story 3-Story 9-Story 20-Story 

Buildings Located in Los Angeles 

Elastic Period (sec.) 1.01 2.24 3.74 1.02 2.21 3.65 

Yield Point Disp. (in.) 2.6 8.0 11.7 2.7 7.7 11.1 

Yield Point Accel. (g) 0.26 0.16 0.09 0.26 0.162 0.085 

Ultimate Point Disp. (in.) 7.5 23 33 8.1 26 44 

Ultimate Point Accel. (g) 0.37 0.23 0.12 0.40 0.27 0.167 

Buildings Located in Seattle 

Elastic Period (sec.) 1.36 3.06 3.46 1.30 3.06 3.52 

Yield Point Disp. (in.) 3.3 7.9 15.0 3.0 7.9 15.5 

Yield Point Accel. (g) 0.18 0.09 0.13 0.18 0.086 0.128 

Ultimate Point Disp. (in.) 9.3 22 43 12.0 25 48 

Ultimate Point Accel. (g) 0.26 0.12 0.18 0.36 0.14 0.198 

Buildings Located in Boston 

Elastic Period (sec.) 1.97 3.30 3.15 1.62 3.17 2.97 

Yield Point Disp. (in.) 2.2 5.8 8.9 3.6 8.0 15.8 

Yield Point Accel. (g) 0.058 0.054 0.091 0.140 0.082 0.183 

Ultimate Point Disp. (in.) 7.1 20 33 10.2 29 47 

Ultimate Point Accel. (g) 0.093 0.095 0.167 0.198 0.150 0.274 

B.5.2 Structural Response 

In the HAZUS methodology, structural response to ground motion is estimated based on 

elastic system properties modified using “effective” stiffness and damping properties of the 

structure to simulate inelastic response. Effective stiffness properties are based on secant 

stiffness at each displacement and effective damping is based on combined viscous and hysteretic 

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measures of dissipated energy, assuming cyclic response of the structure to the given 

displacement. Effective damping greater than 5% of critical is then used to reduce spectral 

demand, in a manner similar to that followed in ATC-40 (ATC, 1997). 

Figure B-3 illustrates the process of developing an inelastic response (demand) spectrum 

from the 5%-damped elastic response (input) spectrum. The demand spectrum is based on 

elastic response divided by amplitude-dependent damping reduction factors (i.e., RA at periods of 

constant acceleration and RV at periods of constant velocity). In Figure B-3, the demand 

spectrum intersects the building’s capacity curve at the point of peak building response (i.e., 

spectral displacement, D, and spectral acceleration, A). The amount of spectrum reduction 

typically increases for buildings that have reached yield and that dissipate hysteretic energy 

during cyclic response. 

Spectral Displacement 

Spectral Acceleration (g’s) 

5%-Damped Response Spectrum 

FVS1 /T 

SSFA 

SSFA /RA 

Building Capacity Curve 

D 

A 

Demand Spectrum 

Area 

FVS1 /TRV 

Figure B-3 Example Demand Spectrum Construction and Calculation of Peak Response 

Point (D, A) 

Spectrum reduction factors are functions of the effective damping beff of the building as 

defined by Equations B-8 and B-9: 

RA = 2.12 (3.21 - 0.68ln( b eff )) (B-8) 

RV = 1.65 (2.31 - 0.41ln( beff )) (B-9) 

Effective damping beff is defined as the total energy dissipated by the building during peak 

earthquake response and is the sum of an elastic damping term bE and a hysteretic damping term 

bH associated with post-yield, inelastic response: 

beff = b E + b H (B-10) 

B-11
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The elastic damping term bE is assumed to be constant (i.e., amplitude independent) and 

represents response at, or just below, the yield point. For most steel moment-frame (WSMF) 

buildings the value of the elastic damping term should be taken as 5% of critical, assuming 

nonstructural components (e.g., cladding) help dampen the structure. The value of the elastic 

damping term should be taken as 3% of critical for bare steel frames or WSMF buildings with 

limited nonstructural damping. 

The hysteretic damping term bH is dependent on the amplitude of post-yield response and is 

based on the area enclosed by the hysteresis loop at peak building displacement D and 

acceleration A as shown in Figure B-3. Hysteretic damping bH is defined in Equation B-11: 

� Area � bH = k � � (B-11) Ł 2p DA ł 

where: Area is the area enclosed by the hysteresis loop, as defined by a symmetrical pushpull of the building capacity curve up to peak positive and negative 

displacements, – D, assuming no degradation of components, 

D is the peak displacement response of the capacity curve,

A is the peak acceleration response at the peak displacement, D

k is a degradation factor that defines the fraction of the Area used to determine

hysteretic damping. 

The k (kappa) factor in Equation B-11 reduces the amount of hysteretic damping as a 

function of anticipated structure performance (e.g., connection condition) and shaking duration, 

to simulate degradation (e.g., pinching) of the hysteresis loop during cyclic response. Shaking 

duration is described qualitatively as either short, moderate or long, and is assumed to be 

primarily a function of earthquake magnitude, although proximity to fault rupture can also 

influence the duration of the level of shaking that is most crucial to building damage. For 

example, ground shaking close to the zone of fault rupture can be strong, but typically contains 

only a few strong pulses. Values of the degradation factor for typical WSMF buildings are 

suggested in Table B-6. 

Table B-6 Values of the Degradation Factor k for Typical WSMF Buildings 

Connection 

Condition 

Peak Response Amplitude and Post-Yield Shaking Duration 

At One-Half 

Yield 

At or Below 

Yield 

Post-Yield Shaking Duration 

Short Moderate Long 

Post-Northridge 1.0 1.0 1.0 0.9 0.7 

Pre-Northridge 1.0 0.9 0.8 0.5 0.3 

Damaged 1.0 0.7 0.6 0.3 0.1 

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As shown in Figure B-3, peak building displacement D is determined by the intersection of 

the capacity curve and the demand spectrum. The intersection requires either a graphical 

solution or a (spreadsheet) calculation that evaluates the area of the hysteresis loop as a function 

of amplitude. Alternatively, the target displacement of Section 3.4.5.3.1, divided by the 

modification factor a2 calculated in accordance with Equation B-5 may used to estimate peak 

nonlinear spectral displacement of the building. In this case, the effective fundamental mode 

period, Te, should be taken as equal to elastic fundamental-mode period Ti and the values of the 

coefficients C1, R, C2 and C3 in Section 3.4.5 should be consistent with structural properties and 

the actual amount of nonlinear response corresponding to the target displacement. 

B.5.3 Structure Fragility 

Building fragility curves are lognormal functions that describe the probability of reaching, or 

exceeding, structural damage states, given deterministic (median) estimates of spectral 

displacement. These curves take into account the variability and uncertainty associated with 

structural response prediction, capacity curve properties, damage states and ground shaking. The 

fragility curves distribute damage among the Slight, Moderate, Extensive and Complete damage 

states. For any given value of spectral response, discrete damage-state probabilities are 

calculated as the difference of the cumulative probabilities of reaching, or exceeding, successive 

damage states. Discrete damage-state probabilities are used as inputs to the calculation of 

building-related losses. 

Each fragility curve is defined by a median value of building response (i.e., spectral 

displacement) that corresponds to the threshold of that damage state and by the uncertainty 

associated with that damage state. The conditional probability of being in, or exceeding, a 

particular damage state ds, given the spectral displacement Sd (or other seismic demand 

parameter), is defined by Equation B-12: 

P[ds Sd ]= Ö 

Ø

Œ 1 ln�

� Sd ��

œ

ø 

(B-12) Œ b ds Ł

� S

ˆ 

d ,ds ł

�

ß

œ o 

where: ˆSd,ds is the median value of spectral displacement at which the building reaches the 

threshold of damage state, ds, 

bds is the standard deviation of the natural logarithm of spectral displacement for 

damage state ds, and 

F is the standard normal cumulative distribution function. 

Development of damage-state medians requires users to: 

• select specific values of maximum interstory drift of the structure that best represent the 

threshold of each of the discrete damage states (consistent with the descriptions of damage 

states provided in Section B.3), and 

• convert damage-state threshold values (e.g., maximum interstory drift) to spectral 

displacement coordinates (i.e., same coordinates as those of the capacity curve). 

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Default values of maximum interstory drift that may be used for typical steel moment-frame 

buildings are provided in Table B-7. These values of drift are consistent with observations of 

damage and loss that occurred in the 1994 Northridge earthquake (pre-Northridge connection 

conditions) and with the interstory drift criteria of Section 3.6 for post-Northridge connection 

conditions. The values of drift given in Table B-7 do not necessarily reflect thresholds of 

damage states of buildings with significant plan or height irregularity. Buildings with a 

significant irregularity would be expected to have substantially smaller values of drift defining 

the thresholds of damage states. 

Table B-7 Maximum Interstory Drift Values Defining Damage-State Thresholds of 

Typical WSMF Buildings 

Connection Condition, Building Height and 

Location 

Structural Damage State 

Slight Moderate Extensive Complete 

Pre-Northridge – All Heights/Locations 0.01 0.015 0.025 0.04 

Post-Northridge – 3-Story – Los Angeles 0.01 0.02 0.040 0.100 

Post-Northridge – 9-Story – Los Angeles 0.01 0.02 0.040 0.080 

Post-Northridge – 20-Story – Los Angeles 0.01 0.02 0.040 0.060 

Post-Northridge – 3-Story – Seattle 0.01 0.0175 0.030 0.080 

Post-Northridge – 9-Story – Seattle 0.01 0.0175 0.030 0.060 

Post-Northridge – 20-Story – Seattle 0.01 0.0175 0.030 0.050 

Post-Northridge – All Heights – Boston 0.01 0.015 0.025 0.04 

Conversion of maximum interstory drift to damage-state medians is based on the building 

height and other factors and the following equation: 

a 2 DdsH ˆ R S d,ds = (B-13)

a3a4,ds 

where: ˆSd ds = median spectral displacement value of damage state, ds (inches) , 

Dds = maximum interstory drift ratio at the threshold of damage state ds, determined 

by user (e.g., typical building values of Table B-8) 

HR = height of building at the roof level (inches) 

a2 = pushover modal factor from Equation B-4 or Equation B-5 

a3 = higher-mode factor (Equation B-14) 

a4,ds = mode-shape factor (Equation B-15) 

The higher-mode factor, a3, is the ratio of interstory drift due to all modes of vibration to the 

interstory drift of the fundamental (pushover) mode at the story with maximum fundamentalmode drift. The value of the higher mode factor may be determined by explicit calculation (e.g., 

ratio of peak drift values of response history and pushover analyses), or may be approximated 

based on the number of stories, N, and the following formula: 

B-14
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a 3 @ N 0.14 £ 1.5 (B-14) 

The mode-shape factor, a4,ds, is the ratio of maximum fundamental-mode (pushover-mode) 

interstory drift to the average pushover-mode interstory drift (i.e., average drift over all stories). 

Maximum pushover-mode interstory drift is the value of drift of those stories contributing to the 

damage state of interest. For tall buildings with Slight structural damage of a localized nature, 

maximum pushover-mode interstory drift is simply the drift of the story with the most 

displacement. As the extent of the damage increases (with damage state) or the building height 

decreases, or both, the difference between maximum pushover-mode interstory drift and average 

pushover-mode interstory drift decreases. The value of the mode-shape factor is 1.0 for 

Complete damage, since damage would typically be pervasive throughout the building. The 

value of the mode-shape factor may be determined directly from the shape of the pushover mode 

or may be approximated based on the number of stories, N, the following formula: 

a4,ds @ N 0.10 (B-15) 

Limits of a4,S £ 1.5 for Slight damage, a4,M £ 1.25 for Moderate damage, a4,E £ 1.1 for 

Extensive damage, and a4,C £ 1.0 for Complete damage are suggested. 

Lognormal standard deviation (b) values describe the total uncertainty inherent in the 

fragility-curve damage states. Three primary sources contribute to the total uncertainty of any 

given state, namely, the uncertainty bC associated with the capacity curve, the uncertainty bD 

associated with the demand spectrum, and the uncertainty bT,ds associated with the discrete 

threshold of each damage state. Since the demand spectrum is dependent on building capacity, a 

convolution process is required to combine their respective contributions to total uncertainty. To 

avoid this rather complex calculation, the Procedures for Developing HAZUS-Compatible 

Building-Specific Damage and Loss Functions (Kircher, 1999) provides pre-calculated values of 

total damage-state uncertainty for different values of capacity, demand and damage state 

variability. Users may refer to this document when developing values of damage-state 

uncertainty or use the b values given in Table B-8 for typical steel moment-frame (WSMF) 

buildings. 

Table B-8 Structural Damage-State Variability (b) Factors of Typical WSMF Buildings 

Building Location Pre-Northridge Connections Post-Northridge Connections 

3-Story 9-Story 20-Story 3-Story 9-Story 20-Story 

Los Angeles 0.90 0.85 0.80 0.70 0.65 0.60 

Seattle 0.95 0.90 0.85 0.75 0.70 0.65 

Boston 0.95 0.90 0.85 

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Commentary: The structural damage state uncertainty factors b given in Table 

B-8 include a large, dominant contribution to the total variability from the 

variability associated with ground shaking demand. A large amount of ground 

shaking variability is appropriate when the fragility functions are to be used to 

estimate damage and loss for a scenario earthquake characterized by median 

predictions of ground shaking. Ground shaking uncertainty accounts for the 

inherent differences between actual and median predictions of ground shaking. 

The structural damage state uncertainty factors b given in Table B-8 would not be 

appropriate for estimating damage when ground shaking is actually known, or for 

estimating probabilistic losses that include ground shaking variability directly in 

the hazard function. 

B.5.4 Loss Functions 

Loss functions convert damage to loss by taking the sum over all four damage states of the 

products of the probability that a building will be damaged within a given damage state 

multiplied by the expected loss given that the damage state is experienced. In the case of 

economic loss, the expected losses can be normalized by dividing by the total replacement value 

to obtain an estimate of the mean loss ratio. 

As discussed in Section B.4, users are expected to provide economic loss data in terms of the 

value of the building (structure), and the costs and associated construction time that would be 

required to repair Slight, Moderate and Extensive damage. These loss parameters would most 

appropriately be based on estimated costs of repair schemes developed to correct Slight, 

Moderate and Extensive damage, as predicted by a performance evaluation (pushover analysis) 

of the structure. Alternatively, default economic loss ratios are provided at the end of this section 

for typical steel moment-frame (WSMF) buildings. 

Repair and replacement costs are the expected dollar costs (per square foot) that would be 

required to repair (or replace) damaged structural elements. In general, the cost of the structural 

system (and related repairs) will vary based on building occupancy (for example, hospital 

structures cost more per square foot then industrial buildings). 

Commentary: Some consideration should be given to prevailing codes and 

ordinances that would govern the repair work. Do prevailing regulations require 

strengthening as well as repair? 

Replacement value is the preferred measure of direct economic loss, although 

other measures could be used, such as loss of market value. Market value would, 

in general, produce entirely different loss estimates. For example, an older 

building of no special importance or historical significance is to be vacated and 

completely renovated, but instead an earthquake occurs and destroys the 

structure. Should economic loss be based on the replacement value (e.g., cost of a 

new building of comparable size and function), on the near zero value of the 

existing building, or on the market value of the building (which would also 

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include value of the land)? These types of question are crucial to the estimation 

of economic loss, but are beyond the scope of this section. For steel momentframe (WSMF) buildings, economic loss functions used here are based on repair 

and replacement value of the structure, consistent with HAZUS methodology. 

Table B-9 provides mean structural repair costs (loss ratios and corresponding loss rates) for 

damage states of typical WSMF buildings. These rates are based on a number of assumptions. 

First, typical WSMF buildings are assumed to have a total replacement value of $125/sq. ft. and 

the structure is assumed to be worth 20% of total building value ($25/sq. ft.). 

Inspection costs of 5% of the cost of the structural system are included in the loss ratios and 

loss rates for buildings with Pre-Northridge connection conditions. The 5% value is based on an 

assumed inspection cost of $1,500 per connection and the assumption that on average about onehalf of the connections of these types of buildings would be inspected following an earthquake. 

The cost of repair of damaged connections is assumed to be $20,000 per connection. On the 

basis of this amount, the cost of repairing all connections would be about one and one-half times 

the cost of a new structural system. 

The cost of repair of Slight damage to buildings with Post-Northridge connection conditions 

is assumed to be zero on the basis that, for example, minor distortion of flanges, or other 

incidental structural damage would not require repair. The cost of repair of Moderate and 

Extensive structural damage of typical WSMF buildings is assumed to be 10% and 50% of the 

value of the structural system. However, the actual repair cost of a specific building could be 

very different due, for example, to the building’s configuration, and the repair’s interference with 

nonstructural systems and finishes. 

Table B-9 Mean Structural Loss Ratios and Rates of Typical WSMF Buildings 

Building Connection Condition Structural Damage State 

Slight Moderate Extensive Complete 

Mean Structural Loss Ratio (Repair Cost / Replacement Cost) 

Pre-Northridge 8% 20% 80% 100% 

Post-Northridge 0% 10% 50% 100% 

Mean Structural Loss Rates (Dollars per Square Foot) 

Pre-Northridge $2.00 $5.00 $20.00 $25.00 

Post-Northridge $0.00 $2.50 $12.50 $25.00 

Repair time is the time required for cleanup and construction to repair or replace damage to 

the structural system. Recovery time is the time required to make repairs, considering, for 

example, delays in decision-making, financing, and inspection, and typically takes much longer 

than the actual time of repair. Loss of function is the time that the facility is not available for use 

and is typically less than repair (recovery) time. Loss of function is less than repair time due to 

temporary solutions, such as the use of alternative space, or simply because buildings with Slight 

or Moderate damage can remain partially or fully operational while repairs are made. Table B-10 

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provides time for cleanup and construction, and loss of function multipliers for typical steel 

moment-frame (WSMF) buildings (mixed occupancy). The loss-of-function multipliers 

represent the fraction of the repair time for each damage state that the building would not be 

functional. 

Table B-10 Cleanup and Construction Time and Loss-of-Function Multipliers for 

Typical WSMF Buildings 

Building Connection Condition and Height Structural Damage State 

Slight Moderate Extensive Complete 

Mean Time of Repairs in Days (Cleanup and Construction) 

Pre-Northridge – 3-Story 5 30 90 180 

Post-Northridge – 3-Story 0 20 90 180 

Pre-Northridge – 9-Story 10 50 180 360 

Post-Northridge – 9-Story 0 40 180 360 

Pre-Northridge – 20-Story 15 75 240 480 

Post-Northridge – 20-Story 0 60 240 480 

Loss-of-Function Multipliers (Fraction of Building Cleanup and Construction Time) 

All Buildings 0.0 0.1 0.3 1.0 

Commentary: The values given in Table B-10 are based on the default values of 

HAZUS adjusted for building height (size) and include time required for 

inspection of WSMF buildings with pre-Northridge connection conditions. 

HAZUS cleanup and repair times and the fractions of repair time that the 

building will not be functional vary widely, depending on the occupancy of the 

building. Values given in Table B-10 are considered appropriate for most 

commercial office buildings. In contrast to HAZUS default values, Slight 

structural damage was assumed to have no impact on building function (loss-offunction multiplier is equal to 0.0, in all cases), since structural inspections and 

repair of connections can typically be made while the building is in operation. 

The loss-of-function multiplier for Complete structural damage is 1.0, and 

assumes that the building is closed and that alternative space is not available. 

B.6 Example Loss Estimates 

This section develops example estimates of losses for typical 9-story Los Angeles buildings, 

designed to conform to the 1994 Uniform Building Code. Three building types are considered: 

(1) buildings with pre-Northridge connection conditions, (2) buildings with post-Northridge 

connection conditions, and (3) buildings with damaged pre-Northridge connection conditions. 

The example considers three levels of earthquake ground shaking that represent the Maximum 

Considered Earthquake (MCE), the Design Earthquake (DE) and one-half of the DE (1⁄2 DE) for 

regions of high seismicity (e.g., Los Angeles). The example first estimates peak building 

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response (spectral displacement) as the intersection of building capacity curves and earthquake 

demand spectra. Building fragility damage and loss functions are then developed using default 

parameters of typical 9-story building properties provided in previous sections. Finally, mean 

building loss functions are developed as a function of building spectral displacement that 

illustrate a range of losses for MCE, DE, 1⁄2 DE, and other levels of spectral demand. 

Commentary: The user is expected to have available an estimate of scenario 

earthquake ground shaking at the building site. Such an estimate may be 

obtained from site-specific hazard studies or from the 1997 USGS/NEHRP 

spectral contour design maps. For this example, 5%-damped response spectra 

were developed from the spectral contour maps representing a typical Los 

Angeles stiff soil site (Soil Profile Type D), not near an active fault. MCE ground 

shaking represents a sufficiently large magnitude event of long shaking duration 

that its approximate return period is between 1,000 to 2,500 years. The DE and 

1⁄2 DE represent ground shaking of a large magnitude event and moderate 

magnitude event, respectively with approximate return periods of 500 and 100 

years, respectively. Most of the steel moment-frame (WSMF) buildings damaged 

by the 1994 Northridge earthquake felt ground shaking that ranged between the 

1⁄2 DE and DE levels illustrated in this example. 

Figure B-4 shows the 5%-damped spectrum of the 1⁄2 DE, the capacity curves of buildings 

with pre-Northridge and post-Northridge connection conditions (solid symbols), and the demand 

curves of buildings with pre-Northridge, post-Northridge and damaged pre-Northridge 

connection conditions (open or shaded symbols). 

0 

0.1 

0.2 

0.3 

0.4 

0.5 

0.6 

0 5 10 15 20 25 30 

Spectral Displacement (inches) 

Spectral Acceleration (g) 

5%-damped spectrum (1/2 DE) 

Pre-Northridge capacity 

Pre-Northridge demand 

Damaged Pre-Northridge demand 

Post-Northridge capacity 

Post-Northridge demand 

Figure B-4 Demand and Capacity of Typical 9-Story WSMF Buildings – Ground 

Shaking of 1⁄2 the Design Earthquake 

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The properties of the capacity curves are based on the yield and ultimate control points given 

in Table B-5. The demand spectra were constructed from the 5%-damped spectrum as described 

in Section B.5.2. The intersection points of demand and capacity curves indicate that spectral 

displacement of the building is about 6.5 inches for each building type. Figures B-5 and B-6 

repeat the process and illustrate the determination of building spectral displacement for Design 

Earthquake (DE) and Maximum Considered Earthquake (MCE) ground shaking, respectively. 

0 

0.1 

0.2 

0.3 

0.4 

0.5 

0.6 

0 5 10 15 20 25 30 

Spectral Displacement (inches) 

Spectral Acceleration (g) 

5%-damped spectrum (DE) 

Pre-Northridge capacity 

Pre-Northridge demand 

Damaged Pre-Northridge demand 

Post-Northridge capacity 

Post-Northridge demand 

Figure B-5 Demand and Capacity of Typical 9-Story WSMF Buildings – Design 

Earthquake Ground Shaking 

0 

0.1 

0.2 

0.3 

0.4 

0.5 

0.6 

0 5 10 15 20 25 30 

Spectral Displacement (inches) 

Spectral Acceleration (g) 

5%-damped spectrum (MCE) 

Pre-Northridge capacity 

Pre-Northridge demand 

Damaged Pre-Northridge demand 

Post-Northridge capacity 

Post-Northridge demand 

Figure B-6 Demand and Capacity of Typical 9-Story WSMF Buildings – Maximum 

Considered Earthquake Ground Shaking 

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Figure B-6 shows different Maximum Considered Earthquake (MCE) intersection points 

(i.e., different values of building spectral displacement) for the three building types. In 

particular, buildings with damaged pre-Northridge connection conditions are expected to degrade 

more than buildings with undamaged connections during the long duration of (post-yield) ground 

shaking associated with the MCE (k = 0.1, Table B-6) and hence are expected to displace farther. 

Table B-11 provides a summary of the predicted peak building response parameters for each 

of the three earthquake ground shaking levels. Spectral displacement is used later in this section 

to estimate structural damage and loss. Table B-11 shows spectral displacement values 

converted to corresponding estimates of average interstory drift, 1st-mode only, average 

interstory drift including higher modes, and maximum interstory drift including higher modes. 

Average interstory drift applies to all stories over the height of the building; maximum interstory 

drift applies to the story experiencing the most displacement. Estimates of drift are based on the 

height of the building (H = 122 feet) and the factors a2, a3 and a4,ds, defined in Section B.5.3. 

Table B-11 Summary of Peak Response – Typical 9-Story WSMF Buildings 

Peak Response Parameter 

Ground Shaking Level – Connection Condition 

1⁄2 DE DE MCE – Long Duration 

All All Post-NR Pre-NR Damaged 

Spectral Displacement (in.) SD 6.5 13 19.5 22 27.5 

Average Interstory Drift -

1st-Mode (SD/H) x 1/a2 

0.006 0.012 0.018 0.020 0.026 

Average Interstory Drift -

All Modes (SD/H) x 1/a2 x a3 

0.008 0.016 0.025 0.028 0.035 

Maximum Interstory Drift – All 

Modes (SD/H) x 1/a2 x a3 x a4,S 

0.010 0.021 0.031 0.035 0.043 

Figure B-7 illustrates structural fragility curves for the example 9-story steel moment-frame 

(WSMF) Los Angeles buildings with post-Northridge connection conditions. These curves are 

constructed using Equation B-12 and the fragility parameters defined in Section B.5.3. Figure B8 illustrates discrete damage-state probabilities for the same buildings. These curves are 

calculated as the difference in probability between adjacent damage-state fragility curves shown 

in Figure B-7. At each value of spectral displacement, the sum of discrete damage-state 

probabilities is equal to the probability of Slight or greater structural damage and the compliment 

of Slight or greater damage is the probability of no structural damage. The considerable overlap 

of discrete damage-state curves shown in Figure B-8 is a measure of the relatively large 

uncertainty in the prediction of damage and is due primarily to the inherent uncertainty in the 

prediction of ground shaking. 

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0 .0 

0 .2 

0 .4 

0 .6 

0 .8 

1 .0 

0 5 10 15 20 25 30 

Pe ak Building Response - Spectral Displacement (Inches) 

Damage-State Probability 

S light or greater 

Moderate o r greater 

Extensive or greater 

Comp lete 

Figure B-7 Structural Fragility Curves – Typical 9-Story Los Angeles Buildings with 

Post-Northridge Connection Conditions 

0.0 

0.1 

0.2 

0.3 

0.4 

0.5 

0.6 

0 5 10 15 20 25 30 

Peak Building Response - Spectral Displacement (Inches) 

Damage-State Probability 

S light 

Moderate 

Extensive 

Comp lete 

Figure B-8 Discrete Damage-State Probability Curves – Typical 9-Story Los Angeles 

Buildings with Post-Northridge Connection Conditions 

Figure B-9 illustrates mean structural loss rates for the structural system of typical 9-story 

steel moment-frame (WSMF) Los Angeles buildings, expressed as a function of building spectral 

displacement. Structural loss ratios are shown for buildings with pre-Northridge and postNorthridge connection conditions to compare the typical reduction in postearthquake repair cost 

that would be expected for buildings with improved connections. Structural loss rates are the 

same for WSMF buildings with pre-Northridge connection conditions, with or without damage to 

connections, although buildings with damaged connections could, depending on the level and 

duration of ground shaking, experience larger spectral displacement and hence greater loss. 

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0 .0 

0 .2 

0 .4 

0 .6 

0 .8 

1 .0 

0 5 10 15 20 25 30 

Pe ak Building Response - Spectral Displacement (Inches) 

Mean Structural Loss Ratio 

Pos t-No rthridge Connection Conditions 

P re-Northridge Connection Conditions 

Figure B-9 Mean Structural Loss Ratio Curves – Typical 9-Story WSMF Los Angeles 

Buildings 

Mean structural loss rate curves are constructed by first multiplying discrete damage-state 

probabilities, shown in Figure B-8, by the mean structural loss rates given in Table B-9, and then 

by summing the products over all damage states. Multiplying mean loss rates by the cost of the 

structural system produces mean estimates of the repair cost (including inspection cost for 

buildings with pre-Northridge connection conditions). For the typical 9-story buildings, the cost 

off the structural system is assumed to be about $5 million (i.e., 20% x $125/sq. ft. x 200,000 sq. 

ft.). Estimates of mean structural loss are made by finding the loss rate corresponding to the 

spectral displacement of the earthquake of interest (e.g., spectral displacement values given in 

Table B-11). 

The results represent mean (or best) estimates of loss rates (rather than a complete 

distribution of loss), since loss rates represent mean (point estimates) of loss, given damage. 

Considering the rather large variability associated with damage estimates (which would only be 

made larger by considering loss uncertainty), actual loss for any given building could be 

significantly different than the mean estimate. The large uncertainty inherent in the fragility 

curves is reflected in the moderate slope of the curve for structural loss. At lower levels of loss, 

loss function tapers to zero gradually with decrease in building spectral displacement. Fragility 

uncertainty is primarily due to the uncertainty associated with median estimates of ground 

shaking. Actual ground shaking could be significantly higher (or lower) than the median and this 

uncertainty tends to broaden the loss functions, increasing estimates of loss at the low end and 

decreasing estimates of loss at the high end (which is typically beyond Maximum Considered 

Earthquake (MCE) demand). 

The mean structural loss rate curves shown in Figure B-9 are plotted to spectral 

displacements of 30 inches, a displacement corresponding to an extremely rare level of 

earthquake ground shaking. Peak building spectral displacements likely to occur during the life 

of the building would not be expected to exceed the 1⁄2 DE level of ground shaking (i.e., about 6 

inches, or less, of spectral displacement). Figure B-10 is a re-plot of Figure B-9 data to a spectral 

displacement of 10 inches. This figure shows that structural repair (and inspection) costs are 

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likely not to exceed 13% of the cost of the structural system ($650,000 loss) on average during 

the life of the building. A loss of 13% is consistent with structural repair (and inspection) costs 

for steel moment-frame (WSMF) buildings damaged in the 1994 Northridge earthquake. The 

figure also shows that repair costs would likely not exceed 2% ($100,000 loss) on average for 

WSMF buildings with post-Northridge connection conditions. For comparison, a typical real 

estate transaction fee for a 200,000 square foot building, based only on the replacement value of 

the building (i.e., excluding the value of the land), would be in excess of $1 million each time the 

building is sold. 

Figure B-11 illustrates mean functional loss (in days) due to damage of the structural system 

of typical 9-story WSMF Los Angeles buildings, expressed as a function of building spectral 

displacement. Functional loss is shown for buildings with pre-Northridge and post-Northridge 

connection conditions to compare the typical reduction in “downtime” for buildings with 

improved connections. Functional loss is the same for WSMF buildings with pre-Northridge 

connection conditions, regardless of connection damage, although buildings with damage 

connections could, depending on the level and duration of ground shaking, experience larger 

spectral displacement and hence greater loss. 

0.00 

0.05 

0.10 

0.15 

0.20 

0.25 

0 2 4 6 8 10 

Peak Building Response - Spectral Displacement (Inches) 

Mean Structural Loss Rate 

Post-Northridge Connection Conditions 

Pre-Northridge Connection Conditions 

Figure B-10 Mean Structural Loss Rate Curves – Typical 9-Story 

WSMF Los Angeles Buildings 

Mean loss of function curves are constructed by first multiplying discrete damage-state 

probabilities, shown in Figure B-8, by the product of the cleanup and construction time and the 

loss-of-function multipliers of Table B-10. Estimates of mean loss of function are made by 

finding the loss of time corresponding to the spectral displacement of the earthquake of interest 

(e.g., spectral displacement values given in Table B-11). 

Mean loss of function (in days) is the probabilistic combination of short downtime due to 

Slight or Moderate structural damage, and long downtime due to Extensive or Complete 

structural damage. Complete damage is assumed to close the building for about the time it 

would take to build a new one (360 days for a 9-story WSMF building). Since the loss-offunction multipliers are very small for Slight or Moderate damage (repairs can usually be made 

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while the building is in operation), loss function is dominated by the probability of Extensive or 

Complete structural damage that would likely close the building for an extended period of time. 

While mean estimates of loss of function are valid as the average of many buildings, actual 

downtime of specific building could range from no loss of function to long-term building 

closure. It may make more sense for users to convert mean loss of function (in days) to a 

probability of long-term building closure by dividing the mean days of downtime by maximum 

down time associated with Complete structural damage. For example, a building with postNorthridge connection conditions is expected to have about 18 days of downtime due to Design 

Earthquake (DE) ground shaking. Actual downtime would likely be considerably less, provided 

the building did not sustain damage sufficient to warrant long-term closure (e.g., a red tag). In 

this case, the probability of long-term closure is about 5% (i.e., mean loss estimate of 18 days 

divided by 360 days of loss associated with Complete damage). 

0 

30 

60 

90 

120 

150 

180 

210 

240 

0 5 10 15 20 25 30 

Pe ak Building Re sponse - Spect ra l Displacemen t (Inches) 

Mean Loss of Function (D ays) 

Pos t-No rthridge Con nec tion Co nd itions 

Pre-North ridge Co nne ction Conditions 

Figure B-11 Mean Loss of Function Curves – Typical 9-Story WSMF Los Angeles 

Buildings 

B-25
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

REFERENCES, BIBLIOGRAPHY, AND ACRONYMS 

This section contains references, additional bibliography and acronyms that are generally 

common to the set of reports, FEMA-350, FEMA-351, FEMA-352, and FEMA-353. Following 

the regular references are three sections containing ASTM Standards published by the American 

Society for Testing and Materials, West Conshohocken, Pennsylvania and listed numerically, 

AWS Specifications published by the American Welding Society, Miami, Florida, and listed 

numerically, FEMA Reports published by the Federal Emergency Management Agency, 

Washington, DC, and listed by report number, and SAC Reports published by the SAC Joint 

Venture, Sacramento, California, and listed by report number. 

References and Additional Bibliography. 

AISC, 1985, Specification for Structural Joints using ASTM A325 or A490 Bolts, American 

Institute of Steel Construction, Chicago, Illinois. 

AISC, 1989, Manual of Steel Construction, ASD, Ninth Edition, American Institute of Steel 

Construction, Chicago, Illinois. 

AISC, 1993, 1997, Load and Resistance Factor Design Specifications for Structural Steel 

Buildings, American Institute of Steel Construction, Chicago, Illinois. 

AISC, 1994a, Proceedings of the AISC Special Task Committee on the Northridge Earthquake 

Meeting, American Institute of Steel Construction, Chicago, Illinois. 

AISC, 1994b, Northridge Steel Update 1, American Institute of Steel Construction, Chicago, 

Illinois. 

AISC, 1997, Seismic Provisions for Structural Steel Buildings, American Institute of Steel 

Construction, Chicago, Illinois. 

AISC, 1998a, Load and Resistance Factor Design Specifications for Structural Steel Buildings, 

American Institute of Steel Construction, Chicago, Illinois. 

AISC, 1998b, LRFD Manual of Steel Construction, 2nd Edition, American Institute of Steel 

Construction, Chicago, Illinois. 

AISC, 1999, Supplement No. 1 to the 1997 Seismic Provisions for Structural Steel Buildings, 

American Institute of Steel Construction, Chicago, Illinois. 

Allen, J., Partridge, J.E., Richard, R.M., and Radau, S., 1995, “Ductile Connection Designs for 

Welded Steel Moment Frames,” Proceedings, 64th Annual Convention, Structural Engineers 

Association of California, Sacramento, California. 

Anderson, J, Duan, J., Xiao, Y., and Maranian, P., 2000, Improvement of Welded Connections 

Using Fracture Tough Overlays, Report No. SAC/BD-00/20, SAC Joint Venture, 

Sacramento, California. 

ASCE, 1998, ASCE-7 maps, American Society of Civil Engineers, Reston, Virginia. 

ASTM citations: see the list of ASTM Standards on page R-4. 

R-1
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

ATC, 1985, Earthquake Damage Evaluation Data for California, Report ATC-13 , Applied 

Technology Council, Redwood City, California. 

ATC, 1987, Evaluating the Seismic Resistance of Existing Buildings, Report ATC-14, Applied 

Technology Council, Redwood City, California. 

ATC, 1989, Procedures for Postearthquake Safety Evaluations of Buildings, Report ATC-20, 

Applied Technology Council, Redwood City, California. 

ATC, 1992, Guidelines for Cyclic Seismic Testing of Components of Steel Structures, Report 

ATC-24, Applied Technology Council, Redwood City, California. 

ATC, 1995, Addendum to the ATC-20 Postearthquake Building Safety Evaluation Procedures, 

Report ATC-20-2, Applied Technology Council, Redwood City, California. 

ATC, 1997a, Seismic Evaluation and Retrofit of Concrete Buildings, prepared by the Applied 

Technology Council (Report No. ATC-40), for the California Seismic Safety Commission 

(Report No. SSC 96-01), Sacramento, California. 

ATC, 1997b, NEHRP Guidelines for the Seismic Rehabilitation of Buildings, Report No. FEMA273, prepared by the Applied Technology Council for the Building Seismic Safety Council, 

published by the Federal Emergency Management Agency, Washington, DC. 

ATC, 1997c, Commentary to NEHRP Guidelines for the Seismic Rehabilitation of Buildings, 

Report No. FEMA-274, prepared by the Applied Technology Council for the Building 

Seismic Safety Council, published by the Federal Emergency Management Agency, 

Washington, DC. 

Avent, R., 1992, “Designing Heat-Straightening Repairs,” National Steel Construction 

Conference Proceedings, Las Vegas, Nevada. 

AWS citations: see the list of AWS reports, specifications and codes on page R-5. 

Barsom, J.M., 1996, “Steel Properties — Effects of Constraint, Temperature, and Rate of 

Loading,” Proceedings of the 2nd US Seminar, Seismic Design, Evaluation and Retrofit of 

Steel Bridges, San Francisco, Report No. UCB/CEE STEEL-96/09, Dept. of Civil and 

Environmental Engineering, UC Berkeley, pp.115-143. 

Boore, D.M., and Joyner, W.B., 1994, Proceedings of Seminar on New Developments in 

Earthquake Ground-Motion Estimation and Implications for Engineering Design Practice, 

Report ATC-35-1, Applied Technology Council, Redwood City, California, pp 6-1 to 6-41. 

BSSC, 1992, NEHRP Handbook for the Seismic Evaluation of Existing Buildings, developed by 

the Building Seismic Safety Council for the Federal Emergency Management Agency, 

Report FEMA-178, Washington, D.C. 

BSSC, 1997a, 1997 NEHRP Recommended Provisions for Seismic Regulations for New 

Buildings and Other Structures, Part 1 – Provisions, prepared by the Building Seismic 

Safety Council for the Federal Emergency Management Agency (Report No. FEMA-302), 

Washington, DC. 

BSSC, 1997b, 1997 NEHRP Recommended Provisions for Seismic Regulations for New 

Buildings and Other Structures, Part 2 – Commentary, prepared by the Building Seismic 

R-2
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

Safety Council for the Federal Emergency Management Agency (Report No. FEMA-303), 

Washington, DC. 

Campbell, K.W., and Bozorgnia, Y., 1994, “Near-Source Attenuation of Peak Horizontal 

Acceleration from Worldwide Accelerograms Recorded from 1957 to 1993,” Fifth U.S. 

National Conference on Earthquake Engineering, Proceedings, Vol. III, pp 283-292, 

Earthquake Engineering Research Institute, Oakland, California. 

Chi, W.M., Deierlein, G., and Ingraffea, A., 1997, “Finite Element Fracture Mechanics 

Investigation of Welded Beam-Column Connections,” SAC Joint Venture, Report No. 

SAC/BD-97/05. 

FEMA citations: see the list of FEMA reports on page R-6. 

Goel, R.K., and Chopra, A.K., 1997, “Period Formulas for Moment-Resisting Frame Buildings,” 

Journal of Structural Engineering, Vol. 123, No. 11, pp. 1454-1461. 

Gross, J.L., Engelhardt, M.D., Uang, C.M., Kasai, K. and Iwankiw, N.R., 1999, Modification of 

Existing Welded Steel Moment Frame Connections for Seismic Resistance, AISC Design 

Guide Series 12, American Institute of Steel Construction, Chicago, Illinois. 

Grubbs, K., 1997, “The Effect of the Dogbone Connection on the Elastic Stiffness of Steel 

Moment Frames” Masters Thesis, Department of Civil Engineering, University of Texas at 

Austin. 

ICBO, 1988, 1991, and 1997, Uniform Building Code, indicated edition, International 

Conference of Building Officials, Whittier, California. 

ICC, 2000, International Building Code, International Code Council, Falls Church, Virginia. 

Kircher, C.A., Nassar, A.A., Kustu, O. and Holmes, W.T., 1997, “Development of Building 

Damage Functions for Earthquake Loss Estimation,” Earthquake Spectra, Vol. 13, No. 4, 

Earthquake Engineering Research Institute, Oakland, California, pp. 663-682. 

Kircher, C.A., Reitherman, R.K., Whitman, R.V., and Arnold, C., 1997, “Estimation of 

Earthquake Losses to Buildings,” Earthquake Spectra, Vol. 13, No. 4, Earthquake 

Engineering Research Institute, Oakland, California, pp. 703-720. 

Kircher, C.A., 1999, Procedures for Development of HAZUS-Compatible Building-Specific 

Damage and Loss Functions, National Institute of Building Sciences, Washington, D.C. 

Krawinkler, H., Gupta, A., Medina, R. and Luco, N., 2000, Loading Histories for Seismic 

Performance Testing of SMRF Components and Assemblies, Report No. SAC/BD-00/10, 

SAC Joint Venture, Sacramento, California. 

NIBS, 1997a, HAZUS Earthquake Loss Estimation Methodology, Users Manual, National 

Institute of Building Sciences, Washington, DC. 

NIBS, 1997b, HAZUS Earthquake Loss Estimation Methodology, Technical Manual, 3 Volumes. 

National Institute of Building Sciences, Washington, DC. 

RCSC, 1996, Load and Resistance Factor Design: Specification for Structural Joints Using 

ASTM A325 or A490 Bolts, Research Council on Structural Connections. 

R-3
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

Richard, R., Partridge, J.E., Allen, J., and Radau, S., 1995, “Finite Element Analysis and Tests of 

Beam-to-Column Connections,” Modern Steel Construction, Vol. 35, No. 10, pp. 44-47, 

American Institute of Steel Construction, Chicago, Illinois. 

SAC citations: see the list of SAC Joint Venture reports on page R-7. 

Shonafelt, G.O., and Horn, W.B, 1984, Guidelines for Evaluation and Repair of Damaged Steel 

Bridge Members, NCHRP Report 271, prepared by the National Cooperative Highway 

Research Program, for the Transportation Research Board, Washington, DC. 

Wald, D.J., Quitoriano, T.H., Kanamori, H. and Scrivner, C.W., 1998, “Trinet Shakemaps – 

Rapid Generation of Peak Ground Motion and Intensity Maps for Earthquakes in Southern 

California”, SMIP98 Proceedings, California Division of Mines and Geology, Sacramento, 

California. 

Whitman, R., Anagnos, T., Kircher, C., Lagorio, H.J., Lawson, R.S., and Schneider, P., 1997, 

“Development of a National Earthquake Loss-Estimation Methodology,” Earthquake 

Spectra, Vol. 13, No. 4, Earthquake Engineering Research Institute, Oakland, California, 

pp. 643-661. 

Youssef, N.F.G, Bonowitz, D., and Gross, J.L., 1995, A Survey of Steel Moment-Resisting Frame 

Buildings Affected by the 1994 Northridge Earthquake, Report No. NISTR 56254, National 

Institute for Science and Technology, Gaithersburg, Maryland. 

ASTM Standards. 

ASTM Standards are published by the American Society for Testing and Materials, West 

Conshohocken, Pennsylvania, and are listed alphanumerically. 

ASTM, 1997, Standard Test Methods and Definitions for Mechanical Testing of Steel Products 

A6, Supplementary Requirement S5 

A36, Specification for Carbon Structural Steel 

A307-97, Standard Specification for Carbon Steel Bolts and Studs, 60 000 PSI Tensile Strength 

A325, Specification for Structural Bolts, Steel, Heat-Treated, 120/105 ksi Minimum Tensile 

Strength 

A435, Straight Beam Ultrasonic Examination of Steel Plates 

A490, Specification for Heat-Treated Steel Structural Bolts, 150 ksi Minimum Tensile Strength 

A563, Specification for Carbon and Alloy Steel Nuts 

A572, Specification for High-Strength Low-Alloy Columbium-Vanadium Structural Steel 

A370-97a, Standard Test Methods and Definitions for Mechanical Testing of Steel Products 

A898, Straight Beam Ultrasonic Examination of Rolled Steel Structural Shapes 

A913, Specification for High-Strength Low-Alloy Steel Shapes of Structural Quality, Produced 

by Quenching and Self-Tempering Process 

A992, Standard Specification for Steel for Structural Shapes for Use in Building Framing 

R-4
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

E329, Standard Specification for Agencies Engaged in the Testing and/or Inspection of Material 

Used in Construction 

E543, Standard Practice for Agencies Performing Nondestructive Testing 

E548, Standard Guide for General Criteria Used for Evaluating Laboratory Competence 

E994, Standard Guide for Laboratory Accreditation Systems 

E1212, Standard Practice for Establishment and Maintenance of Quality Control Systems for 

Nondestructive Testing Agencies 

E1359, Standard Guide for Surveying Nondestructive Testing Agencies 

F436, Specification for Hardened Steel Washers 

F606-98, Standard Test Methods for Determining the Mechanical Properties of Externally and 

Internally Threaded Fasteners, Washers, and Rivets 

F959, Specification for Compressible-Washer-Type Direct Tension Indicators for Use with 

Structural Fasteners 

F1554, Specification for Anchor Bolts, Steel, 36, 55, and 105 ksi Yield Strength 

F1852, Specification for “Twist-Off” Type Tension Control Structural Bolt/Nut/Washer 

Assemblies, Steel, Heat Treated, 120/105 ksi Minimum Tensile Strength 

AWS Reports, Specifications, and Codes. 

AWS reports are published by the American Welding Society, Miami, Florida, and are listed 

alphanumerically. 

AWS A2.4, Standard Symbols for Welding, Brazing, and Nondestructive Testing 

AWS A4.3, Standard Methods for Determination of the Diffusible Hydrogen Content of 

Martensitic, Bainitic, and Ferritic Steel Weld Metal Produced by Arc Welding 

ANSI/AWS A5.1-91, Specification for Carbon Steel Electrodes for Shielded Metal Arc Welding 

ANSI/AWS A5.18-93, Specification for Carbon Steel Electrodes and Rods for Gas Shielded Arc 

Welding 

ANSI/AWS A5.20-95, Specification for Carbon Steel Electrodes for Flux-Cored Arc Welding 

AWS, 1995, Presidential Task Group Report 

ANSI/AWS A5.5-96, Specification for Low-Alloy Steel Electrodes for Shielded Metal Arc 

Welding 

ANSI/AWS A5.28-96, Specification for Low-Alloy Steel Electrodes and Rods for Gas Shielded 

Arc Welding 

ANSI/AWS A5.23/A5.23M-97, Specification for Low-Alloy Steel Electrodes and Fluxes for 

Submerged Arc Welding 

ANSI/AWS A5.25/A5.25M-97, Specification for Carbon and Low-Alloy Steel Electrodes and 

Fluxes for Electroslag Welding 

R-5
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

ANSI/AWS A5.26/A5.26M-97, Specification for Carbon and Low-Alloy Steel Electrodes for 

Electrogas Welding 

ANSI/AWS A5.32/A5.32M-97, Specification for Welding Shielding Gases 

ANSI/AWS A5.17/A5.17M-97, Specification for Carbon Steel Electrodes and Fluxes for 

Submerged Arc Welding 

ANSI/AWS A5.29-98, Specification for Low-Alloy Steel Electrodes for Flux-Cored Arc Welding 

AWS D1.1-1998, 2000, Structural Welding Code – Steel 

AWS D1.3, Structural Welding Code 

AWS D1.4, Structural Welding Code 

AWS QC1, Standard for AWS Certification of Welding Inspectors 

FEMA Reports. 

FEMA reports are listed by report number. 

FEMA-178, 1992, NEHRP Handbook for the Seismic Evaluation of Existing Buildings, 

developed by the Building Seismic Safety Council for the Federal Emergency Management 

Agency, Washington, DC. 

FEMA-267, 1995, Interim Guidelines, Inspection, Evaluation, Repair, Upgrade and Design of 

Welded Moment Resisting Steel Structures, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Washington, DC. 

FEMA-267A, 1996, Interim Guidelines Advisory No. 1, prepared by the SAC Joint Venture for 

the Federal Emergency Management Agency, Washington, DC. 

FEMA-267B, 1999, Interim Guidelines Advisory No. 2, prepared by the SAC Joint Venture for 

the Federal Emergency Management Agency, Washington, DC. 

FEMA-273, 1997, NEHRP Guidelines for the Seismic Rehabilitation of Buildings, prepared by 

the Applied Technology Council for the Building Seismic Safety Council, published by the 

Federal Emergency Management Agency, Washington, DC. 

FEMA-274, 1997, NEHRP Commentary on the Guidelines for the Seismic Rehabilitation of 

Buildings, prepared by the Applied Technology Council for the Building Seismic Safety 

Council, published by the Federal Emergency Management Agency, Washington, DC. 

FEMA-302, 1997, NEHRP Recommended Provisions for Seismic Regulations for New Buildings 

and Other Structures, Part 1 – Provisions, prepared by the Building Seismic Safety Council 

for the Federal Emergency Management Agency, Washington, DC. 

FEMA-303, 1997, NEHRP Recommended Provisions for Seismic Regulations for New Buildings 

and Other Structures, Part 2 – Commentary, prepared by the Building Seismic Safety 

Council for the Federal Emergency Management Agency, Washington, DC. 

FEMA-310, 1998, Handbook for the Seismic Evaluation of Buildings – A Prestandard, prepared 

by the American Society of Civil Engineers for the Federal Emergency Management 

Agency, Washington, DC. 

R-6
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

FEMA-350, 2000, Recommended Seismic Design Criteria for New Steel Moment-Frame 

Buildings, prepared by the SAC Joint Venture for the Federal Emergency Management 

Agency, Washington, DC. 

FEMA-351, 2000, Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded 

Steel Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

FEMA-352, 2000, Recommended Postearthquake Evaluation and Repair Criteria for Welded 

Steel Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

FEMA-353, 2000, Recommended Specifications and Quality Assurance Guidelines for Steel 

Moment-Frame Construction for Seismic Applications, prepared by the SAC Joint Venture 

for the Federal Emergency Management Agency, Washington, DC. 

FEMA-354, 2000, A Policy Guide to Steel Moment-Frame Construction, prepared by the SAC 

Joint Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355A, 2000, State of the Art Report on Base Metals and Fracture, prepared by the SAC 

Joint Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355B, 2000, State of the Art Report on Welding and Inspection, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355C, 2000, State of the Art Report on Systems Performance of Steel Moment Frames 

Subject to Earthquake Ground Shaking, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

FEMA-355D, 2000, State of the Art Report on Connection Performance, prepared by the SAC 

Joint Venture for the Federal Emergency Management Agency, Washington, DC. 

FEMA-355E, 2000, State of the Art Report on Past Performance of Steel Moment-Frame 

Buildings in Earthquakes, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

FEMA-355F, 2000, State of the Art Report on Performance Prediction and Evaluation of Steel 

Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC Joint Venture Reports. 

SAC Joint Venture reports are listed by report number, except for SAC 2000a through 2000k; 

those entries that do not include a FEMA report number are published by the SAC Joint 

Venture. 

SAC 94-01, 1994, Proceedings of the Invitational Workshop on Steel Seismic Issues, Los 

Angeles, September 1994, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC 94-01, 1994b, Proceedings of the International Workshop on Steel Moment Frames, 

Sacramento, December, 1994, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

R-7
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

SAC 95-01, 1995, Steel Moment Frame Connection Advisory No. 3, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Washington, DC. 

SAC 95-02, 1995, Interim Guidelines: Evaluation, Repair, Modification and Design of Welded 

Steel Moment Frame Structures, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Report No. FEMA-267, Washington, DC. 

SAC 95-03, 1995, Characterization of Ground Motions During the Northridge Earthquake of 

January 17, 1994, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC 95-04, 1995, Analytical and Field Investigations of Buildings Affected by the Northridge 

Earthquake of January 17, 1994, prepared by the SAC Joint Venture for the Federal 

Emergency Management Agency, Washington, DC. 

SAC 95-05, 1995, Parametric Analytic Investigations of Ground Motion and Structural 

Response, Northridge Earthquake of January 17, 1994, prepared by the SAC Joint Venture 

for the Federal Emergency Management Agency, Washington, DC. 

SAC 95-06, 1995, Technical Report: Surveys and Assessment of Damage to Buildings Affected 

by the Northridge Earthquake of January 17, 1994, prepared by the SAC Joint Venture for 

the Federal Emergency Management Agency, Washington, DC. 

SAC 95-07, 1995, Technical Report: Case Studies of Steel Moment-Frame Building 

Performance in the Northridge Earthquake of January 17, 1994, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Washington, DC. 

SAC 95-08, 1995, Experimental Investigations of Materials, Weldments and Nondestructive 

Examination Techniques, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Washington, DC. 

SAC 95-09, 1995, Background Reports: Metallurgy, Fracture Mechanics, Welding, Moment 

Connections and Frame Systems Behavior, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Report No. FEMA-288, Washington, DC. 

SAC 96-01, 1996, Experimental Investigations of Beam-Column Subassemblages, Part 1 and 2, 

prepared by the SAC Joint Venture for the Federal Emergency Management Agency, 

Washington, DC. 

SAC 96-02, 1996, Connection Test Summaries, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Report No. FEMA-289, Washington, DC. 

SAC 96-03, 1997, Interim Guidelines Advisory No. 1 Supplement to FEMA-267 Interim 

Guidelines, prepared by the SAC Joint Venture for the Federal Emergency Management 

Agency, Report No. FEMA-267A, Washington, DC. 

SAC 98-PG, Update on the Seismic Safety of Steel Buildings – A Guide for Policy Makers, 

prepared by the SAC Joint Venture for the Federal Emergency Management Agency, 

Washington, DC. 

SAC 99-01, 1999, Interim Guidelines Advisory No. 2 Supplement to FEMA-267 Interim 

Guidelines, prepared by the SAC Joint Venture, for the Federal Emergency Management 

Agency, Report No. FEMA-267B, Washington, DC. 

R-8
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

SAC, 2000a, Recommended Seismic Design Criteria for New Steel Moment-Frame Buildings, 

prepared by the SAC Joint Venture for the Federal Emergency Management Agency, Report 

No. FEMA-350, Washington, D.C. 

SAC, 2000b, Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded Steel 

Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Report No. FEMA-351, Washington, D.C. 

SAC, 2000c, Recommended Postearthquake Evaluation and Repair Criteria for Welded Steel 

Moment-Frame Buildings, prepared by the SAC Joint Venture for the Federal Emergency 

Management Agency, Report No. FEMA-352, Washington, D.C. 

SAC, 2000d, Recommended Specifications and Quality Assurance Guidelines for Steel MomentFrame Construction for Seismic Applications, prepared by the SAC Joint Venture for the 

Federal Emergency Management Agency, Report No. FEMA-353, Washington, D.C. 

SAC, 2000e, A Policy Guide to Steel Moment-Frame Construction, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-354, 

Washington, D.C. 

SAC, 2000f, State of the Art Report on Base Metals and Fracture, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355A, 

Washington, D.C. 

SAC, 2000g, State of the Art Report on Welding and Inspection, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355B, 

Washington, D.C. 

SAC, 2000h, State of the Art Report on Systems Performance, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355C, 

Washington, D.C. 

SAC, 2000i, State of the Art Report on Connection Performance, prepared by the SAC Joint 

Venture for the Federal Emergency Management Agency, Report No. FEMA-355D, 

Washington, D.C. 

SAC, 2000j, State of the Art Report on Past Performance of Steel Moment-Frame Buildings in 

Earthquakes, prepared by the SAC Joint Venture for the Federal Emergency Management 

Agency, Report No. FEMA-355E, Washington, D.C. 

SAC, 2000k, State of the Art Report on Performance Prediction and Evaluation, prepared by the 

SAC Joint Venture for the Federal Emergency Management Agency, Report No. FEMA355F, Washington, D.C. 

SAC/BD-96/01, Selected Results from the SAC Phase 1 Beam-Column Connection Pre-Test 

Analyses, submissions from B. Maison, K. Kasai, and R. Dexter; and A. Ingraffea and G. 

Deierlein. 

SAC/BD-96/02, Summary Report on SAC Phase 1 - Task 7 Experimental Studies, by C. Roeder 

(a revised version of this document is published in Report No. SAC 96-01; the original is no 

longer available). 

SAC/BD-96/03, Selected Documents from the U.S.-Japan Workshop on Steel Fracture Issues. 

R-9
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

SAC/BD-96/04, Survey of Computer Programs for the Nonlinear Analysis of Steel Moment 

Frame Structures. 

SAC/BD-97/01, Through-Thickness Properties of Structural Steels, by J. Barsom and S. 

Korvink. 

SAC/BD-97/02, Protocol for Fabrication, Inspection, Testing, and Documentation of BeamColumn Connection Tests and Other Experimental Specimens, by P. Clark, K. Frank, H. 

Krawinkler, and R. Shaw. 

SAC/BD-97/03, Proposed Statistical and Reliability Framework for Comparing and Evaluating 

Predictive Models for Evaluation and Design, by Y.-K. Wen. 

SAC/BD-97/04, Development of Ground Motion Time Histories for Phase 2 of the FEMA/SAC 

Steel Project, by. P. Somerville, N. Smith, S. Punyamurthula, and J. Sun. 

SAC/BD-97/05, Finite Element Fracture Mechanics Investigation of Welded Beam-Column 

Connections, by W.-M. Chi, G. Deierlein, and A. Ingraffea. 

SAC/BD-98/01, Strength and Ductility of FR Welded-Bolted Connections, by S. El-Tawil, T. 

Mikesell, E. Vidarsson, and S. K. Kunnath. 

SAC/BD-98/02, Effects of Strain Hardening and Strain Aging on the K-Region of Structural 

Shapes, by J. Barsom and S. Korvink 

SAC/BD-98/03, Implementation Issues for Improved Seismic Design Criteria: Report on the 

Social, Economic, Policy and Political Issues Workshop by L.T. Tobin. 

SAC/BD-99/01, Parametric Study on the Effect of Ground Motion Intensity and Dynamic 

Characteristics on Seismic Demands in Steel Moment Resisting Frames by G. A. MacRae 

SAC/BD-99/01A, Appendix to: Parametric Study on the Effect of Ground Motion Intensity and 

Dynamic Characteristics on Seismic Demands in Steel Moment Resisting Frames by G. A. 

MacRae 

SAC/BD-99/02, Through-Thickness Strength and Ductility of Column Flange in Moment 

Connections by R. Dexter and M. Melendrez. 

SAC/BD-99/03, The Effects of Connection Fractures on Steel Moment Resisting Frame Seismic 

Demands and Safety by C. A. Cornell and N. Luco 

SAC/BD-99/04, Effects of Strength/Toughness Mismatch on Structural and Fracture Behaviors 

in Weldments by P. Dong, T. Kilinski, J. Zhang and F.W. Brust 

SAC/BD-99/05, Assessment of the Reliability of Available NDE Methods for Welded Joint and 

the Development of Improved UT Procedures by G. Gruber and G. Light 

SAC/BD-99/06, Prediction of Seismic Demands for SMRFs with Ductile Connections and 

Elements by A. Gupta and H. Krawinkler 

SAC/BD-99/07, Characterization of the Material Properties of Rolled Sections by T. K. Jaquess 

and K. Frank 

SAC/BD-99/08, Study of the Material Properties of the Web-Flange Intersection of Rolled 

Shapes by K. R. Miller and K. Frank 

R-10
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

SAC/BD-99/09, Investigation of Damage to WSMF Earthquakes other than Northridge by M. 

Phipps 

SAC/BD-99/10, Clarifying the Extent of Northridge Induced Weld Fracturing and Examining 

the Related Issue of UT Reliability by T. Paret 

SAC/BD-99/11, The Impact of Earthquakes on Welded steel Moment Frame Buildings: 

Experience in Past Earthquakes by P. Weinburg and J. Goltz 

SAC/BD-99/12, Assessment of the Benefits of Implementing the New Seismic Design Criteria 

and Inspection Procedures by H. A. Seligson and R. Eguchi 

SAC/BD-99/13, Earthquake Loss Estimation for WSMF Buildings, by C. A. Kircher 

SAC/BD-99/14, Simplified Loss Estimation for Pre-Northridge WSMF Buildings, by B. F. 

Maison and D. Bonowitz 

SAC/BD-99/15, Integrative Analytical Investigations on the Fracture Behavior of Welded 

Moment Resisting Connections, by G. G. Deierlein and W.-M. Chi 

SAC/BD-99/16, Seismic Performance of 3 and 9 Story Partially Restrained Moment Frame 

Buildings, by B. F. Maison and K. Kasai 

SAC/BD-99/17, Effects of Partially-Restrained Connection Stiffness and Strength on Frame 

Seismic Performance, by K. Kasai, B. F. Maison, and A. Mayangarum 

SAC/BD-99/18, Effects of Hysteretic Deterioration Characteristics on Seismic Response of 

Moment Resisting Steel Structures, by F. Naeim, K. Skliros, A. M. Reinhorn and M.V. 

Sivaselvan 

SAC/BD-99/19, Cyclic Instability of Steel Moment Connections with Reduced Beam Section, by 

C.-M. Uang and C.-C. Fan 

SAC/BD-99/20, Local and Lateral-Torsion Buckling of Wide Flange Beams, by L. 

Kwasniewski, B. Stojadinovic, and S. C. Goel 

SAC/BD-99/21, Elastic Models for Predicting Building Performance, by X. Duan and J. C. 

Anderson 

SAC/BD-99/22, Reliability-Based Seismic Performance Evaluation of Steel Frame Buildings 

Using Nonlinear Static Analysis Methods, by G. C. Hart and M. J. Skokan 

SAC/BD-99/23, Failure Analysis of Welded Beam to Column Connections, by J. M. Barsom 

SAC/BD-99/24, Weld Acceptance Criteria for Seismically-Loaded Welded Connections, by W. 

Mohr 

SAC/BD-00/01, Parametric Tests on Unreinforced Connections, by K.-H. Lee, B. Stojadinovic, 

S. C. Goel, A. G. Margarian, J. Choi, A. Wongkaew, B. P. Reyher, and D.-Y, Lee 

SAC/BD-00/02, Parametric Tests on the Free Flange Connections, by J. Choi, B. Stojadinovic, 

and S. C. Goel 

SAC/BD-00/03, Cyclic Tests on Simple Connections Including Effects of the Slab, by J. Liu and 

A. Astaneh-Asl 

R-11
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

SAC/BD-00/04, Tests on Bolted Connections, by J. Swanson, R. Leon and J. Smallridge 

SAC/BD-00/05, Bolted Flange Plate Connections, by S. P. Schneider and I. Teeraparbwong 

SAC/BD-00/06, Round Robin Testing of Ultrasonic Testing Technicians, by R. E. Shaw, Jr. 

SAC/BD-00/07, Dynamic Tension Tests of Simulated Welded Beam Flange Connections, by J. 

M. Ricles, C. Mao, E. J. Kaufmann, L.-W. Lu, and J. W. Fisher 

SAC/BD-00/08, Design of Steel Moment Frame Model Buildings in Los Angeles, Seattle and 

Boston, by P. Clark 

SAC/BD-00/09, Benchmarking of Analysis Programs for SMRF System Performance Studies, by 

A. G. and H. Krawinkler 

SAC/BD-00/10, Loading Histories for Seismic Performance Testing of SMRF Components and 

Assemblies, by H. Krawinkler, A. Gupta, R. Medina and N. Luco 

SAC/BD-00/11, Development of Improved Post-Earthquake Inspection Procedures for Steel 

Moment Frame Buildings, by P. Clark 

SAC/BD-00/12, Evaluation of the Effect of Welding Procedure on the Mechanical Properties of 

FCAW-S and SMAW Weld Metal Used in the Construction of Seismic Moment Frames, by 

M. Q. Johnson 

SAC/BD-00/13, Preliminary Evaluation of Heat Affected Zone Toughness in Structural Shapes 

Used in the Construction of Seismic Moment Frames, by M. Q. Johnson 

SAC/BD-00/14, Evaluation of Mechanical Properties in Full-Scale Connections and 

Recommended Minimum Weld Toughness for Moment Resisting Frames, by M. Q. Johnson, 

W. Mohr, and J. Barsom 

SAC/BD-00/15, Simplified Design Models for Predicting the Seismic Performance of Steel 

Moment Frame Connections, by C. Roeder, R.G. Coons, and M. Hoit 

SAC/BD-00/16, SAC Phase 2 Test Plan, by C. Roeder 

SAC/ BD-00/17, Behavior and Design of Radius-Cut, Reduced Beam Section Connections, by 

M. Engelhardt, G. Fry, S. Johns, M. Venti, and S. Holliday 

SAC/BD-00/18, Test of a Free Flange Connection with a Composite Floor Slab, by M. Venti 

and M. Engelhardt 

SAC/BD-00/19, Cyclic Testing of a Free Flange Moment Connection by C. Gilton, B. Chi, and 

C. M. Uang 

SAC/BD-00/20, Improvement of Welded Connections Using Fracture Tough Overlays, by James 

Anderson, J. Duan, P. Maranian, and Y. Xiao 

SAC/BD-00/21, Cyclic Testing of Bolted Moment End-Plate Connections, by T. Murray and E. 

Sumner 

SAC/BD-00/22, Cyclic Response of RBS Moment Connections: Loading Sequence and Lateral 

Bracing Effects, by Q.S. Yu, C. Gilton, and C. M. Uang 

R-12
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings References, Bibliography, and Acronyms 

SAC/BD-00/23, Cyclic Response of RBS Moment Connections: Weak Axis Configuration and 

Deep Column Effects, by C. Gilton, B. Chi, and C. M. Uang 

SAC/BD-00/24, Development and Evaluation of Improved Details for Ductile Welded 

Unreinforced Flange Connections, by J.M. Ricles, C. Mao, L.W. Lu, and J. Fisher 

SAC/BD-00/25, Performance Prediction and Evaluation of Steel Special Moment Frames for 

Seismic Loads, by K. Lee and D. A. Foutch 

SAC/BD-00/26, Performance Prediction and Evaluation of Low Ductility Steel Moment Frames 

for Seismic Loads, by S. Yun and D. A. Foutch 

SAC/BD-00/27, Steel Moment Resisting Connections Reinforced with Cover and Flange Plates, 

by T. Kim, A.S. Whittaker, V.V. Bertero, and A.S.J. Gilani 

SAC/BD-00/28, Failure of a Column K-Area Fracture, by J.M. Barsom and J.V. Pellegrino 

SAC/BD-00/29, Inspection Technology Workshop, by R. E. Shaw, Jr. 

Acronyms. 

A, acceleration response

ACAG, air carbon arc gouging

ACIL, American Council of Independent

Laboratories 

AISC, American Institute for Steel 

Construction 

ANSI, American National Standards Institute 

API, American Petroleum Institute 

ASNT, American Society for Nondestructive 

Testing 

ASTM, American Society for Testing and 

Materials 

ATC, Applied Technology Council 

A2LA, American Association for Laboratory 

Accreditation 

AWS, American Welding Society 

BB, Bolted Bracket (connection) 

BFP, Bolted Flange Plates (connection) 

BOCA, Building Officials and Code 

Administrators 

BSEP, Bolted Stiffened End Plate 

(connection) 

BUEP, Bolted Unstiffened End Plate 

(connection) 

CAC-A, air carbon arc cutting 

CAWI, Certified Associate Welding Inspector 

CJP, complete joint penetration (weld) 

CP, Collapse Prevention (performance level) 

CUREe, California Universities for Research 

in Earthquake Engineering 

CVN, Charpy V-notch 

CWI, Certified Welding Inspector 

D, displacement response 

DST, Double Split Tee (connection) 

DTI, Direct Tension Indicator 

EGW, electrogas welding 

ELF, equivalent lateral force 

ESW, electroslag welding 

FCAW-S, flux-cored arc welding – selfshielded 

FCAW-G, flux-cored arc welding – gasshielded 

FEMA, Federal Emergency Management 

Agency 

FF, Free Flange (connection) 

FR, fully restrained (connection) 

GMAW, gas metal arc welding 

GTAW, gas tungsten arc welding 

HAZ, heat-affected zone 

IBC, International Building Code 

ICBO, International Conference of Building 

Officials 

ICC, International Code Council 

IMF, Intermediate Moment Frame 

IO, Immediate Occupancy (performance 

level) 

R-13
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

References, Bibliography, and Acronyms Steel Moment-Frame Buildings 

ISO, International Standardization 

Organization 

IWURF, Improved Welded Unreinforced 

Flange (connection) 

L, longitudinal 

LDP, Linear Dynamic Procedure 

LRFD, load and resistance-factor design 

LS, Life Safety (performance level) 

LSP, Linear Static Procedure 

MCE, Maximum Considered Earthquake 

MMI, Modified Mercalli Intensity 

MRS, modal response spectrum 

MRSF, steel moment frame 

MT, magnetic particle testing 

NBC, National Building Code 

NDE, nondestructive examination 

NDP, Nonlinear Dynamic Procedure 

NDT, nondestructive testing 

NEHRP, National Earthquake Hazard 

Reduction Program 

NES, National Evaluation Services 

NSP, Nonlinear Static Procedure 

NVLAP, National Volunteer Laboratory 

Accreditation Program 

OMF, Ordinary Moment Frame 

PGA, peak ground acceleration 

PJP, partial joint penetration (weld) 

PIDR, pseudo interstory drift ratio 

PQR, Performance Qualification Record 

PR, partially restrained (connection) 

PT, liquid dye penetrant testing 

PWHT, postweld heat treatment 

PZ, panel zone 

QA, quality assurance 

QC, quality control 

QCP, Quality Control Plan, Quality 

Certification Program 

RBS, Reduced Beam Section (connection) 

RCSC, Research Council for Structural 

Connections 

RT, radiographic testing 

SAC, the SAC Joint Venture; a partnership of 

the Structural Engineers Association of 

California, the Applied Technology 

Council, and California Universities for 

Research in Earthquake Engineering 

SAW, submerged arc welding

SBC, Standard Building Code

SBCCI, Southern Building Code Congress

International 

SCWI, Senior Certified Welding Inspector 

SEAOC, Structural Engineers Association of 

California 

SFRS, seismic-force-resisting system 

SMAW, shielded metal arc welding 

SMF, Special Moment Frame 

SP, Side Plate (connection) 

SUG, Seismic Use Group 

SW, Slotted Web (connection) 

T, transverse 

TIGW, tungsten inert gas welding 

UBC, Uniform Building Code 

UT, ultrasonic testing 

VI, visual inspection 

WBH, Welded Bottom Haunch (connection) 

WCPF, Welded Cover Plate Flange 

(connection) 

WFP, Welded Flange Plate (connection) 

WPQR, Welding Performance Qualification 

Record 

WPS, Welding Procedure Specification 

WSMF, welded steel moment frame 

WT, Welded Top Haunch (connection) 

WTBH, Welded Top and Bottom Haunch 

(connection) 

WUF-B, Welded Unreinforced Flanges – 

Bolted Web (connection) 

WUF-W, Welded Unreinforced Flanges – 

Welded Web (connection) 

R-14
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings SAC Project Participants 

SAC Phase II Project Participants 

FEMA Project Officer FEMA Technical Advisor 

Michael Mahoney Robert D. Hanson

Federal Emergency Management Agency Federal Emergency Management Agency

500 C St. SW, Room 404 DFO Room 353

Washington, DC 20472 P.O. Box 6020

Pasadena, CA 91102-6020 

Joint Venture Management Committee (JVMC) 

William T. Holmes, Chair

Rutherford and Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

Edwin T. Huston

Smith & Huston, Inc.

8618 Roosevelt Way NE

Seattle, WA 98115

Robert Reitherman

California Universities for Research in

Earthquake Engineering 

1301 South 46th St. 

Richmond, CA 94804 

Christopher Rojahn

Applied Technology Council

555 Twin Dolphin Dr., Suite 550

Redwood City, CA 94065

Arthur E. Ross

Cole/Yee/Shubert & Associates

2500 Venture Oaks Way, Suite 100

Sacramento, CA 95833

Robin Shepherd

Earthquake Damage Analysis Corporation

40585 Lakeview Drive, Suite 1B

P.O. Box 1967

Big Bear Lake, CA 92315

Project Management Committee (PMC) 

Stephen A. Mahin, Project Manager

Pacific Earthquake Engr. Research Center

University of California

Berkeley, CA 94720

Ronald O. Hamburger, Project Director for

Project Development

EQE International

1111 Broadway, 10th Floor

Oakland, CA 94607-5500

James O. Malley, Project Director for

Topical Investigations

Degenkolb Engineers

225 Bush St., Suite 1000

San Francisco, CA 94104-1737

William T. Holmes, JVMC

Rutherford and Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

Christopher Rojahn, JVMC

Applied Technology Council

555 Twin Dolphin Dr., Suite 550

Redwood City, CA 94065

Robin Shepherd, JVMC

Earthquake Damage Analysis Corporation

40585 Lakeview Drive, Suite 1B

P.O. Box 1967

Big Bear Lake, CA 92315

S-1
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

SAC Project Participants Steel Moment-Frame Buildings 

Peter W. Clark, Technical Assistant to PMC

SAC Steel Project Technical Office

1301 South 46th St.

Richmond, CA 94804

Project Administration 

Allen Paul Goldstein, Project Administrator Lori Campbell, Assistant to the Project 

A.P. Goldstein Associates

1621B 13th Street

Sacramento, CA 95814

Lee Adler

Structural Engineers Association of

California 

1730 I Street, Ste. 240 

Sacramento, CA 95814 

Administrator 

1621 B 13th Street 

Sacramento, CA 95628 

Project Oversight Committee (POC) 

William J. Hall, Chair

3105 Valley Brook Dr.

Champaign, IL 61821

Shirin Ader

International Conference of Building

Officials 

5360 Workman Mill Rd. 

Whittier, CA 90601-2298 

John M. Barsom

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

Roger Ferch

Herrick Corporation

7021 Koll Center Parkway

P.O Box 9125

Pleasanton, CA 94566-9125

Theodore V. Galambos

University of Minnesota

122 CE Building, 500 Pillsbury Dr. SE

Minnneapolis, MN 55455

John L. Gross

National Institute of Stds. & Technology

Building and Fire Research Lab,

Building 226, Room B158

Gaithersburg, MD 20899

James R. Harris

J.R. Harris and Co.

1580 Lincoln St., Suite 550

Denver, CO 80203-1509

Richard Holguin

520 Kathryn Ct.

Nipomo, CA 93444

Nestor Iwankiw

American Institute of Steel Construction

One East Wacker Dr., Suite 3100

Chicago, IL 60601-2001

Roy Johnston

Brandow & Johnston Associates

1600 West 3rd St.

Los Angeles, CA 90017

S-2
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings SAC Project Participants 

Leonard Joseph

Thornton-Tomassetti Engineers

641 6th Ave., 7th Floor

New York, NY 10011

Duane K. Miller

The Lincoln Electric Company

22801 St. Clair Ave.

Cleveland, OH 44117-1194

John Theiss

EQE/Theiss Engineers

1848 Lackland Hills Parkway

St. Louis, MO 63146-3572

John H. Wiggins

J.H. Wiggins Company

1650 South Pacific Coast Hwy, Suite 311

Redondo Beach, CA 90277

Team Leaders for Topical Investigations 

Douglas A. Foutch

University of Illinois

MC-250, 205 N. Mathews Ave.

3129 Newmark Civil Engineering Lab

Urbana, IL 61801

Karl H. Frank

University of Texas at Austin

10100 Bornet Rd.

Ferguson Lab, P.R.C. #177

Austin, TX 78758

Matthew Johnson

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

Helmut Krawinkler

Department of Civil Engineering

Stanford University

Stanford, CA 94305

Charles W. Roeder

University of Washington

233-B More Hall FX-10

Dept. of Building and Safety

Seattle, WA 98195-2700

L. Thomas Tobin

Tobin and Associates

134 California Ave.

Mill Valley, CA 94941

Lead Guideline Writers 

John D. Hooper

Skilling Ward Magnusson Barkshire, Inc.

1301 Fifth Avenue, Suite 3200

Seattle, WA 98101-2699

Lawrence D. Reaveley

University of Utah

Civil Engineering Dept.

3220 Merrill Engineering Building

Salt Lake City, UT 84112

Thomas A. Sabol

Englekirk & Sabol Consulting Engineers

P.O. Box 77-D

Los Angeles, CA 90007

C. Mark Saunders

Rutherford & Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

Robert E. Shaw

Steel Structures Technology Center, Inc.

42400 W Nine Mile Road

Novi, MI 48375-4132

Raymond H. R. Tide

Wiss, Janney, Elstner Associates, Inc.

330 Pfingsten Road

Northbrook, IL 60062-2095

S-3
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

SAC Project Participants Steel Moment-Frame Buildings 

C. Allin Cornell, Associate Guideline Writer

Stanford University

Terman Engineering Center

Stanford, CA 94305-4020

Technical Advisory Panel (TAP) for Materials and Fracture 

John M. Barsom, POC

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

Serge Bouchard*

TradeARBED

825 Third Avenue, 35th Floor

New York, NY 10022

Michael F. Engestrom*

Nucor-Yamato Steel

P.O. Box 678

Frederick, MD 21705-0678

Karl H. Frank, Team Leader

University of Texas at Austin

10100 Bornet Rd.

Ferguson Lab, P.R.C. #177

Austin, TX 78758

Nestor Iwankiw*

American Institute of Steel Construction

One East Wacker Dr., Suite 3100

Chicago, IL 60601-2001

Dean C. Krouse*

705 Pine Top Drive

Bethelem, PA 18017

Frederick V. Lawrence

University of Illinois at Urbana-Champaign

205 N. Mathews Ave.

Room 2129 Newmark Lab

Urbana, IL 61801

Robert F. Preece

Preece, Goudie & Associates

100 Bush St., Suite 410

San Francisco, CA 94104

Raymond H. R. Tide, Guideline Writer

Wiss, Janney, Elstner Associates, Inc.

330 Pfingsten Road

Northbrook, IL 60062-2095

TAP for Welding and Inspection 

John M. Barsom

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

John W. Fisher

Lehigh University

117 ATLSS Drive

Bethlehem, PA 18015-4729

J. Ernesto Indacochea

University of Illinois at Chicago

Civil and Materials Engineering (mc 246)

842 West Taylor Street

Chicago, IL 60607

Matthew Johnson, Team Leader

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

S-4
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings SAC Project Participants 

David Long

PDM Strocal, Inc.

2324 Navy Drive

Stockton, CA 95206

Duane K. Miller, POC

The Lincoln Electric Company

22801 St. Clair Ave.

Cleveland, OH 44117-1194

Robert Pyle*

AISC Marketing

10101 South State Street

Sandy, Utah 84070

Douglas Rees-Evans*

Steel Dynamics, Inc.

Structural Mill Division

2601 County Road 700 East

Columbia City, IN 46725

Richard I. Seals

P.O. Box 11327

Berkeley, CA 94712-2327

Robert E. Shaw, Guideline Writer

Steel Structures Technology Center, Inc.

42400 W Nine Mile Road

Novi, MI 48375-4132

TAP for Connection Performance 

Charlie Carter*

American Institute of Steel Construction

One East Wacker Drive, Suite 3100

Chicago, IL 60601-2001

Robert H. Dodds

University of Illinois at Urbana-Champaign

205 N. Mathews Ave.

2129 Newmark Lab

Urbana, IL 61801

Roger Ferch, POC

Herrick Corporation

7021 Koll Center Parkway

P.O Box 9125

Pleasanton, CA 94566-9125

John D. Hooper, Guideline Writer

Skilling Ward Magnusson Barkshire, Inc.

1301 Fifth Avenue, Suite 3200

Seattle, WA 98101-2699

Egor Popov

University of California at Berkeley

Department of Civil and Environmental

Engineering, Davis Hall 

Berkeley, CA 94720 

Steve Powell*

SME Steel Contractors

5955 W. Wells Park Rd.

West Jordan, UT 84088

Charles W. Roeder, Team Leader

University of Washington

233-B More Hall FX-10

Dept. of Building and Safety

Seattle, WA 98195-2700

Stanley T. Rolfe

University of Kansas

Civil Engineering Department

2006 Learned Hall

Lawrence, KS 66045-2225

Rick Wilkinson*

Gayle Manufacturing Company

1455 East Kentucky

Woodland, CA 95695

S-5
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

SAC Project Participants Steel Moment-Frame Buildings 

TAP for System Performance 

Jacques Cattan*

American Institute of Steel Construction

One East Wacker Drive, Suite 3100

Chicago, IL 60601-2001

Gary C. Hart

Hart Consultant Group

The Water Garden, Ste. 670E

2425 Olympic Blvd.

Santa Monica, CA 90404-4030

Y. Henry Huang*

Los Angeles County Dept. of Public Works

900 S. Fremont Avenue, 8th Floor

Alhambra, CA 91803

Helmut Krawinkler, Team Leader

Department of Civil Engineering

Stanford University

Stanford, CA 94305

Dennis Randall*

SME Steel Contractors

5955 West Wells Park Road

West Jordan, UT 84088

Andrei M. Reinhorn

State University of New York at Buffalo

Civil Engineering Department

231 Ketter Hall

Buffalo, NY 14260

Arthur E. Ross, JVMC

Cole/Yee/Shubert & Associates

2500 Venture Oaks Way, Suite 100

Sacramento, CA 95833

C. Mark Saunders, Guideline Writer

Rutherford & Chekene

303 Second St., Suite 800 North

San Francisco, CA 94107

W. Lee Shoemaker*

Metal Building Manufacturers Association

1300 Summer Avenue

Cleveland, OH 44115

John Theiss, POC

EQE/Theiss Engineers

1848 Lackland Hills Parkway

St. Louis, MO 63146-3572

TAP for Performance Prediction and Evaluation 

Vitelmo V. Bertero

University of California at Berkeley

Pacific Earthquake Engr. Research Center

1301 S. 46th St.

Richmond, CA 94804

Bruce R. Ellingwood

Johns Hopkins University

Department of Civil Engineering

3400 N. Charles St.

Baltimore, MD 21218

Douglas A. Foutch, Team Leader

University of Illinois

MC-250, 205 N. Mathews Ave.

3129 Newmark Civil Engineering Lab

Urbana, IL 61801

Theodore V. Galambos, POC

University of Minnesota

122 CE Building, 500 Pillsbury Dr. SE

Minnneapolis, MN 55455

Lawrence G. Griffis

Walter P. Moore & Associates

3131 Eastside, Second Floor

Houston, TX 77098

Edwin T. Huston, JVMC

Smith & Huston, Inc.

8618 Roosevelt Way NE

Seattle, WA 98115

S-6
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings SAC Project Participants 

Harry Martin*

American Iron and Steel Institute

11899 Edgewood Road, Suite G

Auburn, CA 95603

Thomas A. Sabol, Guideline Writer

Englekirk & Sabol Consulting Engineers

Tom Schlafly*

American Institute of Steel Construction

One East Wacker Drive, Suite 3100

Chicago, IL 60601-2001

P.O. Box 77-D

Los Angeles, CA 90007

NormAbrahamson

Pacific Gas & Electric

P.O. Box 770000, MC N4C

San Francisco, CA 94177

C.B. Crouse

URS – Dames and Moore

2025 First Avenue, Suite 500

Seattle, WA 98121

Technical Advisors 

Robert Kennedy

RPK Structural Mechanics Consultants

18971 Villa Terr

Yorba Linda, CA 92886

Social Economic and Policy Panel 

Martha Cox-Nitikman

Building and Owners and Managers

Association, Los Angeles 

700 South Flower, Suite 2325 

Los Angeles, CA 90017 

Karl Deppe

27502 Fawnskin Dr.

Rancho Palos Verdes, CA 90275

Eugene Lecomte

Institute for Business and Home Safety

6 Sheffield Drive

Billerica, MA 01821

James Madison

Attorney at Law, Mediator and Arbitrator

750 Menlo Avenue, Suite 250

Menlo Park, CA 94025

Alan Merson

Morley Builders

2901 28th Street, Suite 100

Santa Monica, CA 90405

Joanne Nigg

University of Delaware

Disaster Research Center

Newark, DE 19716

William Petak

University of Southern California

Lewis Hall, Room 201

650 Childs Way

Los Angeles, CA 90089

Francine Rabinovitz

Hamilton, Rabinovitz and Alschuler

1990 South Bundy Drive, Suite 777

Los Angeles, CA 90025

S-7
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

SAC Project Participants Steel Moment-Frame Buildings 

Dennis Randall

SME Steel Contractors

5955 West Wells Park Road

West Jordan, UT 84088

David Ratterman

Stites and Harbison

400 West Market St., Suite 1800

Louisville, KY 40202-3352

L. Thomas Tobin, Panel Coordinator

134 California Ave.

Mill Valley, CA 94941

Stephen Toth

TIAA-CREF

730 Third Avenue

New York, NY 10017-3206

John H. Wiggins, POC

J.H. Wiggins Company

1650 South Pacific Coast Hwy, Suite 311

Redondo Beach, CA 90277

Performance of Steel Buildings in Past Earthquakes Subcontractors 

David Bonowitz

887 Bush, No. 610

San Francisco, CA 94108

Peter Clark

SAC Steel Project Technical Office

1301 South 46th St.

Richmond, CA 94804

Michael Durkin

Michael Durkin & Associates

22955 Leanora Dr.

Woodland Hills, CA 91367

James Goltz

California Institute of Technology

Office of Earthquake Programs

Mail Code 252-21

Pasadena, CA 91125

Bruce Maison

7309 Lynn Ave

Elcerrito, CA 94530

Peter Maranian

Brandow & Johnston Associates

1660 West Third Street

Los Angeles, CA 90017

Terrence Paret

Wiss Janney Elstner Associates, Inc.

2200 Powell St. Suite 925

Emeryville, CA 94602

Maryann Phipps

Degenkolb Engineers

225 Bush Street, Suite 1000

San Francisco, CA 94104

Allan Porush

Dames & Moore

911 Wilshire Blvd., Suite 700

Los Angeles, CA 90017

Access Current Knowledge Subcontractors 

David Bonowitz Stephen Liu

887 Bush , No. 610 Colorado School of Mines

San Francisco, CA 94108 Mathematics and Computer Science

Department 

Golden, CO 80401 

S-8
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings SAC Project Participants 

Materials and Fracture Subcontractors 

Robert Dexter Karl H. Frank

University of Minnesota University of Texas at Austin

122 Civil Engineering Building 10100 Bornet Rd.

500 Pillsbury Drive SE Ferguson Lab, P.R.C. #177

Minneapolis, MN 55455-0116 Austin, TX 78758

Welding and Inspection Subcontractors 

Pingsha Dong / Tom Kilinski Glenn M. Light / George Gruber 

Center for Welded Structures Research Southwest Research Institute 

Battelle Memorial Institute

501 King Avenue

Columbus, OH 43201-2693

Matthew Johnson

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

6220 Culebra Road, P. O. Drawer 28510

San Antonio, TX 78228-0510

William C. Mohr

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

Connection Performance Subcontractors 

Gregory Deierlein

Stanford University

Terman Engineering Center

Department of Civil and Enviromental Engr.

Stanford, CA 94305-4020

Charles W. Roeder

University of Washington

233-B More Hall FX-10

Seattle, WA 98195-2700

Sherif El-Tawil / Sashi Kunnath

University of Central Florida

Civil and Environmental Engr. Department

Orlando, FL. 32816-2450

Anthony Ingraffea

Cornell University

School of Civil Engineering

363 Hollister Hall

Ithaca, NY 14853

System Performance Subcontractors 

Paul Somerville

Woodward-Clyde Federal Services

566 El Dorado St., Suite 100

Pasadena, CA 91101-2560

Farzad Naeim

John A. Martin & Associates

1212 S. Flower Ave.

Los Angeles, CA 90015

Andrei M. Reinhorn

State University of New York at Buffalo

Civil Engineering Department

231 Ketter Hall

Buffalo, NY 14260

C. Allin Cornell

Stanford University

Terman Engineering Center

Stanford, CA 94305-4020

S-9
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

SAC Project Participants Steel Moment-Frame Buildings 

Helmut Krawinkler

Dept. of Civil Engineering

Stanford University

Stanford, CA 94305

Gregory MacRae

University of Washington

Civil Engineering Department

Seattle, WA 98195-2700

Kazuhiko Kasai

Tokyo Institute of Technology

Structural Engineering Research Center

Nagatsuta, Midori-Ku

Yokohama 226-8503, JAPAN

Bruce F. Maison

7309 Lynn Avenue

El Cerrito, CA 94530

Performance Prediction and Evaluation Subcontractors 

James Anderson

University of Southern California

Civil Engineering Department

Los Angeles, CA 90089-2531

Douglas A. Foutch

University of Illinois

MC-250, 205 N. Mathews Ave.

3129 Newmark Civil Engineering Lab

Urbana, IL 61801

Gary C. Hart

Department of Civil and Environmental

Engineering 

University of California 

Los Angeles, CA 90095 

Y.K. Wen

University of Illinois

3129 Newmark Civil Engineering Lab

205 N. Mathews Ave.

Urbana, IL 61801

Testing Subcontractors 

Subhash Goel / Bozidar Stojadinovic

University of Michigan

Civil Engineering Department

Ann Arbor, MI 48109

Roberto Leon

Georgia Institute of Technology

School of Civil & Environmental Engr.

790 Atlantic Ave.

Atlanta, GA 30332-0355

Vitelmo V. Bertero / Andrew Whittaker

UC Berkeley

Pacific Earthquake Engr. Research Center

1301 S. 46th St.

Richmond, CA 94804

Thomas Murray

Virginia Tech, Dept. of Civil Engineering

200 Patton Hall

Blacksburg, VA 24061

James M. Ricles / Le-Wu Lu

Lehigh University

c/o ATLSS Center

117 ATLSS Drive, H Building

Bethlehem, PA 18015-4729

John M. Barsom

Barsom Consulting, Ltd.

1316 Murray Ave, Suite 300

Pittsburgh, PA 15217

S-10
Recommended Seismic Evaluation and Upgrade 

Criteria for Existing Welded FEMA-351 

Steel Moment-Frame Buildings SAC Project Participants 

Hassan Astaneh

University of California at Berkeley

Dept. of Civil and Environmental Engr.

781 Davis Hall

Berkeley, CA 94720

Michael Engelhardt

University of Texas at Austin

Ferguson Laboratory

10100 Burnet Road, Building 177

Austin, TX 78712-1076

Gary T. Fry

Texas A&M University

Department of Civil Engineering

Constructed Facilities Division, CE/TTI

Building, Room 710D 

College Station, TX 77843-3136 

Chia-Ming Uang

University of California at San Diego

Dept. of AMES, Division of Structural Engr.

409 University Center

La Jolla, California 92093-0085

Stephen Schneider

University of Ilinois at Urbana-Champaign

3106 Newmark Civil Engr. Lab, MC-250

205 N. Mathews Avenue

Urbana, IL 61801

Matthew Johnson

Edison Welding Institute

1250 Arthur E. Adams Drive

Columbus, OH 43221

James Anderson

University of Southern California

Civil Engineering Department

Los Angeles, CA 90089-2531

Bozidar Stojadinovic

Dept. of Civil & Environmental Engr.

University of California

Berkeley, CA 94720

Inspection Procedure Consultants 

Thomas Albert

Digiray Corporation

2235 Omega Road, No. 3

San Ramon, CA 94583

Randal Fong

Automated Inspection Systems, Inc.

4861 Sunrise Drive, Suite 101

Martinez, CA 94553

Andre Lamarre

R.D Tech, Inc.

1200 St. Jean Baptiste, Suite 120

Quebec City, Quebec, Canada G2ZE 5E8

Glenn Light

Southwest Research Institute

6220 Culebra Road

San Antonio, TX 78228

Andrey Mishin

AS & E High Energy Systems

330 Keller Street, Building 101

Santa Clara, CA 95054

Robert Shaw

Steel Structures Technology Center, Inc.

42400 W. Nine Mile Road

Novi, MI 48375-4132

Carlos Ventura

Dept of Civil Engineering

University of British Columbia

2324 Main Hall

Vancouver, BC, Canada V6T 1Z4

S-11
Recommended Seismic Evaluation and Upgrade 

FEMA-351 Criteria for Existing Welded 

SAC Project Participants Steel Moment-Frame Buildings 

Guideline Trial Applications Subcontractors 

John Hopper Lawrence Novak

Skilling Ward Magnusson Barkshire, Inc. Skidmore, Owings, and Merrill

1301 Fifth Avenue, Suite 320 

Seattle WA 98101-2699 

Leonard Joseph 

Thornton-Tomassetti Engineers 

641 6th Avenue, 7th Floor 

New York, NY 10011 

224 S. Michigan Ave, Suite 1000 

Chicago, IL 60604 

Maryann Phipps 

Degenkolb Engineers 

225 Bush Street, Suite 1000 

San Francisco, CA 94104 

Economic and Social Impact Study Subcontractors 

Ronald Eguchi

EQE Engineering and Design

300 Commerce Dr., Ste. 200

Irvine, CA 92602

Martin Gordon / Peter Morris

Adamson Associates

170 Columbus Avenue

San Francisco, CA 94133

Richard Henige

Lemessurier Consultants Inc.

675 Massachusetts Ave.

Cambridge, MA 02139-3309

Charles Kircher

Charles Kircher & Associates

1121 San Antonio Road, Suite D-202

Palo Alto, CA 94303

Lizandro Mercado

Brandow & Johnston Associates

1600 West 3rd St.

Los Angeles, CA 90017

Greg Schindler

KPFF Consulting Engineers

1201 3rd Ave.

Seattle, WA 98101-3000

Report Production and Administrative Services 

A. Gerald Brady, Technical Editor Carol Cameron, Publications Coordinator

Patricia A. Mork, Administrative Asst. Ericka Holmon, Admin. Assistant

Peter N. Mork, Computer Specialist California Universities for Research in

Bernadette A. Mosby, Operations Admin. Earthquake Engineering

Michelle S. Schwartzbach, Pub. Specialist 1301 S. 46th Street

Applied Technology Council Richmond, CA 94804

555 Twin Dolphin Drive, Suite 550

Redwood City, CA 94065

*indicates industrial or organizational contact representative 

S-12