NBS SPECIAL PUBLICATION 470

Vau of "

U.S. DEPARTMENT OF COMMERCE / National Bureau of Standards

WIND AND SEISMIC EFFECTS

Proceedings of the

Seventh Joint UJNR

Panel Conference
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WIND AND SEISMIC EFFECTS

Proceedings of the Seventh Joint

Panel Conference of the U.S.-Japan

Cooperative Program in

Natural Resources

May 20-23, 1975

Tokyo, Japan

H.S. Lew, Editor

Center for Building Technology

Institute for Applied Technology

National Bureau of Standards

Washington, D.C. 20234

if W %

U.S. DEPARTMENT OF COMMERCE, Juanita M. Kreps, Secretary

Dr. Betsy Ancker-Johnson, Assistant Secretary for Science and Technology

\v S> NATIONAL BUREAU OF STANDARDS. Ernest Ambler, Acting Director

Issued April 1977
Library of Congress Cataloging in Publication Data

United States-Japan Cooperative Program in Natural Resources. Panel on Wind and Seismic Effects. Wind and seismic effects.

(National Bureau of Standards special publication ; 470)

Supt. of Docs, no.: C13. 10:470

1. Wind-pressure-Congresses. 2. Earthquakes and buildingCongresses. I. Lew, Hai Sang. II. Title. III. Series: United States. National Bureau of Standards. Special publication ; 470. QC100U57 no. 470 [TA654.5] 602'.ls [624'. 176] 77-608015

National Bureau of Standards Special Publication 470

Nat. Bur. Stand. (U.S.), Spec. Publ. 470, 513 pages (Apr. 1977)

CODEN: XNBSAV

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WASHINGTON: 1977

For sale by the Superintendent of Documents, U.S. Government Printing Office, Washington, D.C. 20402

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PREFACE

The Seventh Joint Meeting of the U.S. - Japan Panel on Wind and Seismic Effects was held in Tokyo, Japan on May 20-23, 1975. This panel is one of the twenty panels in the U.S.

Japan Cooperative Program in Natural Resources (UJNR) . The UJNR was established in 1964 by

the U.S. - Japan Cabinet-level Committee on Trade and Economic Affairs. The purpose of the

UJNR is to exchange scientific and technological information which will be mutually beneficial to the economics and welfare of both countries. Accordingly, the purpose of the

annual joint meeting of this panel is to exchange technical information on the latest research and development activities within governmental agencies of both countries in the area

of wind and seismic effects.

The proceedings include the program, the formal resolutions, and the technical papers

presented at the Joint Meeting. The papers were presented in the respective language of

each country. The texts of the papers, all of which were prepared in English, have been

edited. The illustrations were reproduced from the working documents used at the Joint

Meeting. The formal resolutions were drafted at the closing session of the Joint Meeting

and adopted unanimously by the panels of both countries.

Pages of the technical papers are numbered with a prefix corresponding to the Theme

number. The texts are consecutively numbered in each theme.

H. S. Lew, Secretary

U.S. Panel on Wind and

Seismic Effects
SI Conversion Units

In view of present accepted practice in this technological area, U.S. customary units

of measurements have been used throughout this report. It should be noted that the U.S. is

a signatory to the General Conference on Weights and Measures which gave official status to

the metric SI system of units in 1960. Conversion factors for units in this report are:

Customary Unit International

(SI) , UNIT

Conversion

Approximate

Length inch (in)

foot (ft)

meter (m)

meter (m)

1 in=0. 0254m*

1 ft=0. 3048m*

Force

Pressure

Stress

Energy

Torque

or Bending

Moment

pound (Ibf)

kilogram (kgf)

pound per square

inch (psi)

Kip per square

inch (ksi)

inch-pound (in- Ibf)

foot-pound (ft-lbf)

pound-inch

pound- foot (Ibf-ft)

newton (N)

newton (N)

newton/meter^

newton/meter

^

joule (J)

joule (J)

newton-meter (N-m)

newton-meter (N-m)

1 lbf=4.448N

1 kgf=9.807N

1 psi=6895N/m'^

1 ksi=5895xlO^N/m^

1 in- Ibf=0.1130 J

1 ft- Ibf=1.3558 J

1 lbf-in=0.1130 N-m

1 Ibf- ft=l. 3558 N-m

Weight

or

Mass

Unit Weight

Velocity

Acceleration

pound (Ibf)

pound per cubic foot

(pcf)

foot per second

(ft/sec)

foot per second per

2 second (ft/sec )

kilogram (kg) 1 lb=0.4536 kg

kilogram per cubic

meter (kg/m^)

meter per second

(m/s)

meter per second per

2 second (m/s )

1 pcf=16. 018 kg/m"

1 fps=0.3048 m/s

1 ft/sec^=0.3048 m/s^

Meter may be subdivided. A centimeter (cm) is 1/100 m and a millimeter (mm) is 1/1000 m.

* Exact
CONTENTS

Preface iii

SI Conversion iv

Abstract vii

Program of the Joint Meeting ix

Opening Session ix

List of Members xiii

Formal Resolutions xvii

Themes and Technical Papers I-l

THEME I: CHARACTERISTICS OF STRONG WIND

Present Status of Wind Characteristics

Kiyonide Takeuchi .. I-l

A Re-examination of Hurricane Camille

Arnold R. Hull 1-18

Fire-Tornadoes and Its Maximum Wind Speed

S. Soma and K. Suda 1-28

A Research Project on the Wind Flow Around Tall Buildings

Tatsuo Murota and Kiyoshi Nakano 1-42

THEME II: RESPONSE OF FULL-SCALE STRUCTURES TO WIND ACTION

Wind Engineering Research Program Supported by the National Science

Foundation

Michael P. Gaus II-l

Study of the Wind Pressure and the Response of Roof Corners

Tatsuo Murota and Mitsuo Makahara 11-11

On the Wind Response of the Kanmon Bridge

T. Okubo, N. Narita, and K. Yokoyama 11-20

Luling, Louisana Cable-Stayed Bridge Wind Tunnel Section Model Tests

Richard H. Gade, Walter Podolny, Jr. and Harold R. Bosch 11-47

THEME III: GEOLOGICAL DISTRIBUTION OF SEISMIC ACTIVITY

Regional Distribution of Earthquake Risk in Japan

S. Hattori, Y. Kitagawa and T. Santo III-l

Quantification of Seismicity

Teutomu Terashima and Tetsuo Santo III-18

THEME IV: MAINTENANCE OF STRONG MOTION ACCELEROGRAPHS AND DATA PROCESSING

Maintenance of the Strong Motion Accelerograph and Data Processing of

Records Obtained

Eiichi Kuribayashi, Hajime Tsuchida and Makoto Watabe IV-1

The United States Strong-Motion Network: Field Operation

Richard P. Maley IV-35

Strong-Motion Data Management

Christopher Rojahn IV- 5

3

V
THEME V: STRONG EARTHQUAKE MOTIONS AND GROUND FAILURES

Brief Review on Liquefaction During Earthquakes in Japan

Eiichi Kuribayashi and Fumio Tatsuoka V-1

Vibration Test on Settlement of Submerged Sand Layer

K. Sawada and Y. Koga V-16

THEME VI: RESPONSE OF HYDRAULIC AND BUILDING STRUCTURES TO SEISMIC FORCES

Study on Earthquake Response of Structures by Considering Non-Deterministic

Variables

Yutaka Yamazaki and Yasunori Koizume VI-1

Least Weight Structures for Threshold Frequencies

R.D. McConnell VI-22

Ductile Shear Walls in Earthquake-Resistant Multistory Buildings

Mark Fintel VI- 3 3 Surveillance of Corps of Engineers Structures in Earthquake-prone Area

Keith O. O'Donnell VI-47

School and Hospital Construction in California

John F. Meehan VI-64

Improved Earthquake Resistive Design and Construction of Single Fan^ily.

Residental Dwellings

G. Robert Fuller VI-74

THEME VII: ASEISMIC CONSIDERATIONS FOR VESSELS

Dynamic Tests of Structures for Oil Tanks and Nuclear Power Plants

Seiichi Inaba VII-1

Sheet Pile Foundation and its Structural Characteristics against

Horizontal Loads

Kenji Kawakami, Tadyoshi Okubo, Keiichi Komada and Michio Okahara VII-7

THEME VIII: RECENT REVISIONS OF DESIGN STANDARDS ON WIND AND SEISMIC EFFECTS

A Review on Specifications of Earthquake Resistant Design for Highway

Bridges - 1971

Kenji Kawakami, Eiichi Kuribayashi, Toshio Iwasaki and Yutaku lida VIII-1

Specifications for Earthquake Resistant Design of the Honshu-Shikoku

Bridges (JSCE 1974)

Ishio Kawasaki and Biichi Kuribayashi VIII-31

Recent Revision of Design Standards on Seismic Effect for Port and

Harbour Structures

Satoshi Hayashi, Hajime Tsuchida and Setsuo Noda VIII-42

JSCE Specifications for Earthquake Resistant Design of Submerged Tunnels

(1975)

Eiichi Kuribayashi and Hajime Tsuchida VIII-56

THEME IX: JOINT RESEARCH PROGRAM UTILIZING LARGE SCLAE TESTING FACILITIES

Joint Research Program Utilizing the Large Scale Testing Facilities

(Free Discussion)

Makota Watabe, Masaya Hirosawa and Shinsuke Nakata IX-1

Earthquake Disaster Mitigation: A Joint Research Approach

Charles T. Theil IX-6

vi
THEME X: TECHNOLOGICAL ASSISTANCE TO DEVELOPING COUNTRIES

High Wind Study in the Philippines

Noel J. Raufaste X-1

Survey on Seismology and Earthquake Engineering in India, Iran and Turkey

Makoto Watabe, Hideo Tokuhiro and Masakazu Shinozuka X-12

vii
ABSTRACT

The Seventh Joint Meeting of the U.S. - Japan Panel on Wind and Seismic Effects was held in Tokyo, Japan on May 20-23, 1975. The proceedings of the Joint Meeting include the

program, the formal resolutions, and the technical papers. The subject matter covered in

the papers includes characteristics of strong wind; response of full-scale structures to

wind action; geological distribution of seismic activity; maintenance of strong motion

accelerographes and data processing; strong earthquake motions and ground failures; response

of hydraulic and building structures to seismic forces; aseismic considerations for vessels;

recent revisions of design standards on wind and seismic effects; joint research program

utilizing large scale testing facilities; and technological assistance to developing

countries

.

Key Words: Accelerograph; bridges; buildings; codes; disaster; dynamic analysis; earthquakes; ground failures; hydraulic structures; seismicity; soils; standards;

storage tanks; structural response; winds.

viii
SEVENTH JOINT MEETING

PROGRAM OF THE

U.S. - JAPAN PANEL ON WIND AND SEISMIC EFFECTS

May 20-23, 1975

at

Tokyo , Japan

TUESDAY - May 20

Opening Ceremony (Room HANA NO MA, 6F, Zenkyoren Building)

9:30 A.M. Call to order by Mr. Hiroshi Yoshimura, Head, Planning Division,

Public Works Research Institute

Remarks by Mr. Mitsuo Kikuchi, Engineer General, Ministry of Construction

f Remarks by Mr. William W. Henoch, Acting Counselor for Scientific and

Technological Affairs, U.S. Embassy

Remarks by Mr. Saburo Ueda, Chief, International Division Promotion Bureau,

Science and Technology Agency

Remarks by Mr. Kenji Kawakami, Director, Public Works Research Institute,

Ministry of Construction

Remarks by Dr. E.O. Pfrang, Chief, Structures, Materials & Life Safety

Division Center for Building Technology, lAT, National Bureau of Standards

Introduction of United States Panel Members by U.S. Chairman and Japan

Panel Members by Japanese Chairman

Election of Conference Chairman

Adoption of Agenda

11:00 A.M. Explanation of the 1974 Off Izu-Peninsula Earthquake

11:30 A.M. Lunch

THEME III: Geological Distribution of Seismic Activity

11:00 P.M. Regional Distribution of Earthquake Risk in Japan

S. Hattoru, Y. Kitagawa and T. Scuto

1:25 P.M. Quantification of Seismicity

Tsutomii Terashima, Tetsuo Santo

1:50 P.M. Discussion

2:10 P.M. Recess

THEME IV: Maintenance of Strong Motion

Accelerographs and Data Processing

2:25 P.M. Maintenance of the Strong Motion Accelograph and Data Processing of Records

Obtained

Eiichi Kuribayashi, Hagime Tsuchida and Makoto Watabe

2:50 P.M. The United States Strong-Motion Network: Field Operation

Richard P. Maley

3:15 P.M. Strong-Motion Data Management

Christopher Rojahn: presented by R. Maley

ix
3:40 P.M. Discussion

Courtesy Call (at the office of the Minister of Construction)

4:45 P.M. Courtesy call to Mr. T. Kariya, Minister of Construction

Reception (at the Rose-Room 6F, in Zenkyoren Building)

6:00 P.M. Reception to be hosted by Mr. M. Kikuchi, Engineering General

8:00 P.M. Ministry of Construction

WEDNESDAY - May 21

THEME V: Strong Earthquake Motions

and Ground Failures

9:00 A.M. Brief Review on Liquefaction During Earthquakes in Japan

Eiichi Kuribayashi and Fumiotatsuoka

9:25 A.M. Vibration Test on Settlement of Submerged Sand Layer

S. Sawada and Y. Koga

9:50 A.M. Discussion

10:50 A.M. Recess

THEME VI: Response of Hydraulic and Building

Structures to Seismic Forces

10:25 A.M. Study on Earthquake Response of Structures by Considering Non-Deterministic

Variables

Yutaki Yamazaki and Yasunori Koizimi

10:50 A.M. Least Weight Structures for Threshold Frequencies

Richard D. McConnell «

11:15 A.M. Ductile Shear Walls in Earthquake-Resistant Multistory Buildings

Mark Fintel

11:40 A.M. Discussion

12:00 P.M. Lunch

1:30 P.M. Surveillance of Corps of Engineers Structures in Earthquake-prone Area

Keith O. O'Donnell

1:55 P.M. School and Hospital Construction in California

John F. Meehan; presented by E.O. Pfrang

2:20 P.M. Improved Earthquake Resistive Design and Construction of Single Family

Residential Dwelling

G. Robert Fuller

2:45 P.M. Discussion

3:05 P.M. Recess

X
THEME VII: Recent Revisions of Design Standards

on Wind and Seismic Effects

3:20 P.M. A Review on Specifications of Earthquake Resistant Design for Highway

Bridges - 1971

3:45 P.M. Specifications for Earthquake Resistant Design of the Honshu-Shikoku

Bridge (JSCE-1974)

Ishio Kawasaki and Etichi Kuribayashi

4:10 P.M. Discussion

4:30 P.M. Recent Revision of Design Standards on Seismic Effect for Port and Harbour

Structures

Sagoshi Hayashi, Hajime Tsuchida and Setsuo Noda

4:55 P.M. JSCE Specifications for Earthquake Resistant Design of Submerged Tunnels

(1975)

Eiichi Kuribayashi and Hajime Tsuchida

5:20 P.M. Discussion

THURSDAY - May 22

THEME I: Characteristics of Strong Wind

9:00 A.M. Present Status of Wind Characteristics in Japan

Kiyohide Takeuchi

9:25 A.M. A Reexamination of Hurricane Camille

Arnold R. Hull

9:50 A.M. Discussion

10:10 A.M. Recess

10:25 A.M. Fire-Tornadoes and its Maximum Wind Speed

S. Soma and K. Suda

10:50 A.M. High Wind Study in the Philippines

Noel Raufaste

THEME II: Response of Full Scale Structures

To Wind Action

11:15 A.M. Wind Engineering Research Program Supported by the National Science Foundation

Michael P. Gaus; presented by E.O. Pfrang

11:40 A.M. Discussion

12:00 P.M. Lunch

1:30 P.M. A Research Project on the Wind Flow Around Tall Buildings

Tatsuo Murota and Kyoshi Nakano

1:55 P.M. Study of the Wind Pressure and the Response of Flat Roof Corners

Tatsuo Murato

2:20 P.M. Luling Louisiana Cable-stayed Bridge Wind Tunnel Section Model Tests

Richard H. Cade; presented by J. Cooper

Walter Podoling, Jr. and Harold R. Busch

xi
2:45 P.M. Discussion

3:05 P.M. Recess

3:20 P.M. On the Wind Response of the Kanmon Bridge

T. Okubo, N. Narita and K. Yokoyama

3:45 P.M. Newport, Rhode Island Suspension Bridge; Hurricane "Doria" Wind Spectra

Richard H. Gade; presented by J. Cooper

4:10 P.M. Discussion

THEME VII: A Seismic Considerations for Vessels

4:30 P.M. Dynamic Tests of Structures for Oil Tanks and Nuclear Power Plants

Seiichi Inaba

4:55 P.M. Sheet Pile Foundation and its Structural Characteristic against

Horizontal Loads

Kenji Kawakami, Tadyoshi Okubo, Keiichi Komada and Michio Okahara

5:20 P.M. Discussion

FRIDAY - May 23

THEME IX: Joint Research Program Utilizing Large

Scale Testing Facilities

9:00 A.M. Joint Research Program Utilizing the Large Scale Testing Facilities

(Free Discussion)

Makota Watabe, Masaya Hirosawa and Shinsuke Nakata

9:25 A.M. Disaster Mitigation: A Joint Approach

Charles T. Thiel; presented by E.O. Pfrang

THEME X: Technological Assistance to Developing Countries

9:50 A.M. Survey on Seismology and Earthquake Engineering in India, Iran and Turkey

Makoto Watabe, Hideo Tokuhiro and Masekazu Shinozuka

10:15 A.M. Discussion

10:35 A.M. Recess

10:55 A.M. Explanation of Oita Earthquake

12:00 P.M. Lunch

1:30 P.M. Free Discussion

2:00 P.M. Closing Session

3:28 P.M. Lv. Tokyo

xii
U. S. PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

APRIL 1975

Dr. Edward 0. Pfrang, Chairman

Chief, Structures, Materials and

Safety Division

Center for Building Technology, lAT

National Bureau of Standards

Room B368, Bldg. 226

Washington, D.C. 20234

(301) 921-2196

Dr. S. T. Algermissen, Member

Seismicity and Risk Analysis Branch

USGS, R10X4, RB3, Rm. A434

Boulder, Colorado 80302

(303) 234-4014

Dr. Robert D. Borcherdt, Member

USGA

National Center for Earthquake Research

345 Middlefield Road

Menlo Park, California 94025

(415) 323-8111

Dr. George Ericksen, Member

USGS

12201 Sunrise Valley Drive

Reston, Virginia 22092

(703) 860-6553

Dr. Michael P. Gaus, Member

Head, Engineering Mechanics Section

Engineering Division

National Science Foundation

1800 G. Street, N.W.

Washington, D.C. 20550

(202) 632- 5787

Dr. R. Cecil Gentry, Member

Director, National Hurricane Research

Laboratory

NOAA

P. 0. Box 8265, University of Miami Branch

Coral Gables, Florida 33124

(305) 350-4150

Mr. Sigmund Gerber, Member

Director for Construction Standards and

Design for Installation and Logistics

Office of the Secretary of Defense

Room 3E763, The Pentagon

Washington, D.C. 20301

(202) 695-2713

Mr. Jack Healy, Member

Chief, Material Systems and Science Division

Department of the Army

Construction Engineering Research Labs.

P. 0. Box 4005

Champaign, Illinois 61820

(217) 352-6416 (FTS only)

(217) 352-6511 (commercial)

Mr. Arnold R. Hull, Member

Associate Director for Climatology

NOAA-EDS

Washingtin, D.C. 20235

(202) 634-7319

Dr. William B. Joyner, Member

USGS

Office of Earthquake Research

345 Middlefield Road

Menlo Park, California 94025

(415) 323-8111 X2754

Mr. John W. Kaufman, Member

Chief, Atemspheric Dynamics Branch

Aerospace Environment Division

National Aeronautics & Space Administration

Marshall Space Flight Center, Alabama 35812

(205) 453-3103

Mr. James Lefter, Member

Director, Civil Engineering Service

Office of Construction

Veterans Administration

Lafayette Building, Room 507

811 Vermont Avenue, N.W.

Washington, D.C. 20420

(202) 389-2864

Dr. R.B. Matthiesen, Member

Chief, Seismic Engineering Branch

USGS

390 Main Street, Room 7067

San Francisco, California 94105

(415) 556-7768

Mr. Keith O. O'Donnell, Member

Assistant Chief, Structural Branch

Engineering Division, Civil Works

Directorate

Forrestal Building, ATTN: DAEN-CWE-D

Washington, D.C. 20314

(202) 0X3-7311

xiii
Mr. Larry F. Rousch, Member

Public Building Service

General Services Administration

18th & F Streets, N.W.

Washington, D.C. 20405

Dr. Charles Scheffey, Member

Acting Associate Administrator

Federal Highway Administration

Washington, D.C. 20590

(202) 426-2943

Dr. Warren A. Shaw, Member

Head, Civil Engineering Department

Civil Engineering Laboratory

Naval Construction Batallion Center

Port Hueneme, California 93043

(805) 982-5407

Mr. Lawrence C. Shao, Member

Directorate of Licensing

U.S. Atomic Energy Commission

Washington, D.C. 20545

(201) 973-7808

Dr. Charles T. Thiel, Member

Program Manager, Earthquake Engineering

Division of Advanced Technology Applications

National Science Foundation

Washington, D.C. 20550

(202) 632-0648

Mr. William J. Werner, Member

Division of Building Technology and Safety

Department of Housing and Urban Development

Room 4120

451 7 Street, S.W.

Washington, D.C. 20410

(202) 755-5574

ALTERNATES

Dr. H.S. Lew, Secretary and Alternate

Structures, Materials and Life Safety

Division

Center for Building Technology, lAT

National Bureau of Standards

Washington, D.C. 20234

(301) 921-3851

Dr. Richard D. McConnell, Alternate

Veterans Administration

Office of Construction

Washington, D.C. 20420

Dr. John B. Scalzi, Alternate

Division of Advanced Technology and

Applications

National Science Foundation

Washington, D.C. 20550

(202) 632-5970

Mr. F.J. Tamanini, Alternate

Chief, Structures, and Applied Mechanics

Division

HRS-10

Federal Highway Administration

U.S. Department of Transportation

Washington, D.C. 20550

(703) 557-5287

xiv
MEMBERSHIP LIST

Japan Panel

Mr. Kenji Kawakami, Chairman

Director, Public Works Research Institute

Ministry of Construction

2- 28-32 Honkomagome , Bunkyo-ku, Tokyo

Dr. Satoshi Hayashi

Head, Structure Division

Port and Harbor Research Institute,

Ministry of Transport

3- 1-1 Nagase, Yokosuka-shi, Kanagawa-ken

Mr. Seiichi Inaba

Chief, Earthquake Engineering Laboratory

National Research Center for Disaster

Prevention

Science and Technology Agency

4489 Kurihara Aza, Sakuramura,

Niihari-gun, Ibaragi-ken

Dr. Yasunori Koizumi

Director, Building Research Institute

Ministry of Construction

3-28-8 Hyak\inin-cho, Shinjuku-ku, Tokyo

Mr. Keiichi Komada

Chief, Founding Engineering Section

Structure and Bridge Division, Chiba Branch

Public Works Research Institute,

Ministry of Construction

4- 12-52 Anagawa, Chiba-shi, Chiba-ken

Mr. Eiichi Kuribayashi

Chief, Earthquake Engineering Section

Structure and Bridge Division, Chiba Branch

Public Works Research Institute, Ministry

of Construction

4-12-52 Anagawa, Chiba-shi, Chiba-ken

Mr. Tatsuo Murota

Senior Research Officer, Third Research

Division

Building Research Institute

Ministry of Construction

3-28-8 Hyakunin-cho, Shinjuku-ku, Tokyo

Dr. Kiyoshi Nakano

Head, Structural Division, Building

Research Institute

Ministry of Construction

3-28-8 Hyakunin-cho, Shinjuku-ku, Tokyo

Mr. Nobuyuki Narita

Chief, Structure Section

Structure and Bridge Division,

Chiba Branch

Public Works Research Institute,

Ministry of Construction

4-12-52 Anagawa, Chiba-shi, Chiba-ken

Dr. Tadayoshi Okubo

Head, Structure and Bridge Division,

Chiba Branch

Public Works Research Institute,

Ministry of Construction

4-12-52 Anagawa, Chiba-shi, Chiba-ken

Dr. Tetsuo Santo

Head, International Institute of Seismology

and Earthquake Engineering

Building Research Institute, Ministry of

Construction

3- 28-8 Hyakunin-cho, Shinjuku-ku, Tokyo

Mr. Kenkichi Sawada

Chief, Soil Dynamics Section

Construction Method and Equipment Division

Chiba Branch

Px±>lic Works Research Institute, Ministry

of Construction

4- 12-52 Anagawa, Chiba-shi, Chiba-ken

Dr. Seiji Soma

Chief, First Research Section

Physical Meteorology Laboratory

Meteorological Research Institute

Japan Meteorological Agency, Ministry of

Transport

4-35-8 Kita, Koenji, Suginami-ku, Tokyo

Dr. Ken Suda

Director, Meteorolocical Research Institute

Japan Meteorological Agency, Ministry of

Transport

4-35-8 Kita, Koenji, Suginami-ku, Tokyo

Mr. Akira Suwa

Head, Seismo logical Laboratory

Meteorological Research Institute

Japan Meteorological Agency, Ministry of

Transport

4-35-8 Kita, Koenji, Suginami-ku, Tokyo

XV
Dr. Kiyohide Takeuchi

Head, Applied Meteorological Laboratory

Meteorological Research Institute

Japan Meteorological Agency, Ministry of

Transport

4-35-8 Kita Koenji, Suginami-ku, Tokyo

tMr. Hajime Tsuchida

Chief, Earthquake Resistant Structures

Laboratory

Structure Division

Port and Harbor Research Institute,

Ministry of Transportation

3-1-1 Ngase, Yokosuka-shi, Kanagawa-ken

Dr. Makoto Watabe

Chief, International Institute of Seismology

and Earthquake Engineering

Building Research Institute,

Ministry of Construction

3-28-8 Hyakunin-cho, Shinjuku-ku, Tokyo

Secretary-General

Mr. Hiroshi Yoshimura

Head, Planning Division

Public Works Research Institute

Ministry of Construction

2-28-32 Honkomagome , Bunkyo-ku, Tokyo

xvl
RESOLUTIONS OF SEVENTH JOINT MEETING

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS

U.J.N.R.

May 20 - 23, 1975

The following resolutions for future U.S. -Japan activities of the Joint Panel on Wind

and Seismic Effects are hereby proposed:

1. The Seventh Joint Meeting of U.S. -Japan Panel on Wind and Seismic Effects, U.J.N.R.

Program, was extremely fruitful to both countries. Considering the importance of

the technical information exchanged between the two delegations, it is considered

essential that the joint program be continued

2. Due to the continuous nature of the research programs reported, it is recommended

that dissemination of technical reports and state-of-art reports be encouraged on

a continual basis during the coming year

3. In accordance with the resolutions of the Sixth Joint Panel Meeting, the proceedings of the Joint Meeting should be published for the benefit of other

scientific representatives in each country as soon as possible

4. Efforts should be made to exchange, on a timely basis, list of significant strongmotion earthquake and high speed wind recorded data. Clearly defined good quality

digitizable record copies are needed and a procedure for exchanging records should

be established. Records should also include the characteristics of instruments

needed for all processing and analytical procedures

5. Attempts should be made to solicit papers from other official governmental offices

and private technical organizations involved in wind and seismic research. Papers

written by non-panel members may be synopsized in state-of-art reports

6. Promote the exchange of technological information, concerning strong winds and

earthquakes, with developing countries

7. Cooperative research programs including exchange of personnel and equipment

should be undertaken by both governments to address the following problems of

mutual interest:

a) Strong-motion instrumentation arrays, at selected sites throughout the

world,

b) Large-scale testing programs,

c) Repair and retrofit of existing structures, ie; buildings, bridges, dams, etc.

d) Structural performance evaluation,

e) Land use programs for controlling natural hazard effects,

f ) Disaster prevention methods for lifeline systems

.

8. It is suggested that all future papers incorporate the SI Metric System for units

of measurement, contained within parenthesis

9. Future reports should consider an overview of codes and problems of implementation

associated with "nonstructural" and "operational" requirements and subsequently

should be exchanged

xvii
10, More exposure and exchange of innovative techniques for improved seismic design

spectra is considered highly important

11. The date and location for the 8th Joint Panel Meeting on Wind and Seismic Effects

will be Spring 1976 in Washington, D.C. , U.S.A. The specific date, inspection

sites and field trips will be determined by the U.S. Panel with approval by the

Japanese Panel.

A formal expression of appreciation is hereby presented by the U.S. Panel members to

the Japanese delegation for the excellent arrangements, technical exchange and magnificent

hospitality received at this Seventh Joint Panel Meeting in Tokyo, May 1975.

xviii
PRESENT STATUS OF WIND CHARACTERISTICS IN JAPAN

by

Kiyohide Takeuchi

Applied Meteorology Laboratory

Meteorological Research Institute

ABSTRACT

The present status of a study on the wind characteristics in Japan is given herein.

Observational studies constructed from a tower and an array of towers are described in

addition to results obtained from tethered balloons.

The observations obtained from towers in Tokushima Pref . and analysis of the data,

which were made by Shiotani (1972 and 1974) , are detailed and unique. Also other observations from towers located at Tarama Island, Okinawa Pref. are being conducted by Mitsuta

(1974) . Detailed analysis of these observations are also presented.

Finally some model experiments on the local wind, using a tunnel, are also presented.

Key Words: Field Measurements; Model; Towers; Wind; Wind Observation; Wind Tunnel.

I-l
Introduction

The behavior of wind has been under constant study in Japan, and is one of the most

fundamental sxJDjects in meteorology. Meteorological disasters in Japan are caused mainly

by the wind and rain, as induced by typhoons, cyclones and monsoons.

Tall buildings and structures as well as long suspension bridges have recently been

constructed in Japan. Accordingly detailed information of the wind characteristics is

required for their proper design and for maintenance after construction.

However, air pollution caused by various industries and cars has become a serious problem in Japan. Since the wind in the atmospheric boundary layer plays an essential role

in the dispersion of pollutants, a study of the wind, especially the local wind in the industrial and urban areas is quite urgent.

Since the interaction between the air and the earth surface is an essential problem

in meteorology, a study on the wind fluctuation near the surface has been intensively

conducted.

The characteristics of the wind has been investigated in Japan through observations conducted on field towers and through wind tunnel experiments in the laboratory.

Observations made in the field were conducted on single towers and tower arrays in

order to obtain more detailed information on the time and space characteristics of the

wind. In addition to using tethered balloons, pilot balloons, low altitude radiosondes,

low altitude rawin sondes and aircraft were often used for observing the local wind.

The present status of the study on wind characteristics is described in detail herein.

Observation from a Single Tower

Observation from a Tower Near the Mouth of River Naka

In order to obtain data required for the design of a long suspension bridge, since 1964

Shiotani (1975) has studied wind characteristics from a tower located near the mouth of

River Naka, Tokushima Pref. This meteorological tower, 150m in height, is located

approximately 150m away from the beach (see Fig. 1) . The base of the tower is on a low

sand dune covered with pine trees, a few meters in height, but in north north-west directiorv

a marsh extends a distance of 400m. The land surrounding the tower, which is 1 km from

the coast, is under cultivation with intermittant farm houses, trees and small hills.

This meteorological tower is a guyed mast made of steel pipe. The Aerovane-type anemometers were installed 3 m apart from the center of the tower at 15.8, 30.8, 50.8, 80.8, 110.8

and 150.8 m above the ground. The mean wind speed was continuously recorded during a

minute period, at each anemometer level. Instantaneous wind speeds and wind directions were measured at any three levels with pen-writing recorders when the wind speed became high.

In addition, measurement of the vertical direction of the wind was made with two bidirectional vanes. Temperature differences between 150 and 30 m, and 80 and 30 m were measured

with bead- thermistor thermometers.

The wind behavior during a 15 min interval were recorded every 2.5 s when wind speeds

were under 25 m/s, and every 2.0 s when wind speed were over 25 m/s. From these data mean

wind speed U, turbulent velocity O (i.e., r.m.s. of the longitudinal wind fluctuation),
auto correlation, power spectrum S(n) (n: frequency), and other statistical quantities

were obtained. Analyses of these data 197 recordings indicated that the range of wind

speeds was between 20 and 30 m/s, and only 13 data had wind speeds higher than 30 m/s.

Since the characteristics of the high wind are determined mainly by the surface

roughness, the wind data has, therefore, divided into three groups. The first group had a

wind direction 1 (E-ES) - the wind from the sea after passing 150-300 m over the land.

The second. Direction 2 (SSE-WSW) had the wind passing more than 2 km over the delta of

River Naka, which is modified from the sea wind. The third group, Direction 3 (WNW-NNW)

has a wind parallel to the coastal line (see Fig. 1) . The results from these classifications are as follows;

(a) Velocity Fluctuation and Their Magnitudes

The frequency distribution of the fluctuating wind velocities can be expressed

approximately by Gaussian distribution. The intensity of turbulence (a^/U) of the

wind in Direction 2 is higher than that from the other two directions , and it is

recognized that the turbulent intensity increases with increasing roughness.

The intensity of turbulence of the wind at 30.8 m is much higher than that at

the other heights. This means that the rough ground coverage, in the close proximity to the tower, greatly disturbs the wind in the lowest region. The turbulent

intensities of the wind in the range between 50.8 m and 150.8 m are nearly the

same in Directions 1 and 3, and they decrease with increasing heights. However,

they are approximately constant with those heights in Direction 2. In order to

examine the vertical distribution of wind turbulent velocities, their ratios to

those recorded at 80.8 m, i.e. o /a (80), were studied. These results indicate

u u that they are nearly constant with height in Direction 2, while they decrease with

power exponents -0.10 and -0.16 in Directions 1 and 3, respectively

.

(b) Power Spectrum

The relationship between the logarithmic spectrum nS(n) and the logarithm of the wave number log(n/U), was examined. It was noted that of the 261 recordings, 139

recordings had single peak spectrum nS (n) . These single-peak spectra were then

used for further analyses. The shapes of the spectra are quite varied, therefore,

the whole spectra cannot be represented by a single curve. The wave number (n/U) , corresponding to the maximum value of nS(n), is generally m

independent of the wind speed and wind direction. Therefore, the height distribution of (n/U) averaged at each height, was found that the wave nxunber of the m

spectrum peak decreases slightly with increasing height.

Examination of the similarity theory, however, suggests that nS (n) can be expressed

by the factor nz/U, therefore, examination of collected data results in the following empirical formula;

o ft!? (nz/U) = 0.0018 z

.1-3
This indicated that the spectrum peak shifts slightly to a low-frequency with an increase in height, for a height range between 50.8 m and 150.8 m. However, the

accuracy of the formula is not so restrictive such that (n/U) can be considered

m constant with height.

The friction velocity u^ can be estimated from the high-frequency part of the

spectrum. The mean value of O /u. can then be evaluated, giving 2.4 for the case

u * of Direction 2. However, u^ can also be estimated from the logarithmic law of the mean wind speed, which gives a value for the same direction, equal to 2.1.

(c) Correlation Coefficient Between Two Heights

An example of the correlation coefficient R{z^,z^,t) with time lag T between

the two heights and {z^>z^) , is shown in Fig. 2. It should be noted that

the maximum correlation does not occur when x = 0, but occurs when the wind record

of the upper layer is delayed. This means that the phase of the wind gusts ad- vances when in the upper air layer. Now, using Taylor's hypothesis, the space

correlation is as shown in Fig. 3. Examination of this Figure shows that the

scale of the turbulence becomes larger and the phase advances during an increase

in air layer height.

The vertical correlation coefficient R{z^,z^) appears to be independent of the

wind direction and wind speed. Therefore, the correlation coefficient will depend

not only on the height difference (z^-z^) , but on the actual heights z^ and z^.

Examination of the correlation coefficients between two points which have the

same height differences indicates that the largest value in the case where the

two points are in the highest air layer, which means the scale of turbulence in- creases upward. The correlation coefficient is empirically formulated as follows:

R(Z2,Z^) = exp[-0.85(z2''"'^^ - z^^^^)]

The coherence function Y(z2,z^,n) between the two heights z^ and z^ can now be

calculated. The coherence is found empirically and is expressed as:

1/2 [Y(z2,z^,n)] = exp[-kn(z2 - z^)/V^^]

Because the correlation coefficient between the two heights depends on the height

difference, k should be a function of two heights. Also, it has been formed that

k increase as the wind speed increases and the height of the air layer decreases.

Observation at the NHK Kawaguchi Tower

Many observations of the atmohpheric boundary layer have been made by using broadcasting towers in or near the city. For example. Soma (1964) has analyzed the data obtained at the Tokyo Tower and has studied the turbulent structure of high wind over the

urban area.

J-4.
since air pollution has recently become a serious problem in the urban area, various

meteorological observations have often been made. The following will describe such observations taken at the NHK Broadcasting tower near Tokyo.

In 1971 and 1972, Yokoyama and his collaborators installed sonic anemometer-thermometers, to measure the wind fluctuations which are closely associated with diffusion of

pollutants (see Mitsuta, 1966 for the sonic anemometer-thermometer) . The heights of the

anemometers used were 10, 45, 90, 180 and 313 m, respectively. The tower is located in

a suburban residential area where the ground is rather flat. Measured signals were

recorded by long-time magnetic tape recorders and the play-back signals were processed

by an analogue-digital converter and a computer. Various statistical quantities, such as turbulent velocity and momentum flux, were then analyzed. The analyses of these quantities,

indicate the following;

(a) Magnitude of Vertical Fluctuation

For a constant flux layer (i.e., layer up to about 50 m above the ground) it is well

known that the magnitude of vertical fluctuation O is related to the friction velocity w

u^ as follows;

a = Au^cj) (z/L) w * w

The term (f)^ is a universal function of the stability a/L (L: Monin-Obukhov stability

length) and A is constant. In the constant flux layer, the friction velocity u^ is

obtained from the vertical distribution of the mean wind speed and also from the co- variance of wind fluctuation. In a layer higher than the constant flux layer, the equation mentioned above does apply. However, when the local friction velocity obtained

1/2 from the covariance (i.e., u^ = -u' w' ) is used, the equation applies in the higher

layer. The relation between a /u^ and the stability z/L is then studied and d) is w * w

determined as a function of the stability.

(b) Stability and Relation between O and U w

In many cases at height of 45 m, the magnitude of vertical fluctuation is proportional to the mean velocity U. At a height of 313 m, however, it seems that there

exists a linear relation only over the certain value which might be determined by

the stability as shown in Fig. 4.

Observations at the Tower in Okinawa and Iwo-Jima

Kinase et al. (1972) have made wind observations by using the OHK TV-Tower in Okinawa

(165 m high above the ground) and Loran Tower in Iwo-Jima (410 m high). Aerovane-type

anemometers were installed at 6 levels at the OHK TV-Tower and at 8 levels at the Loran

Tower. Since there existed strong electromagnetic fields, special attention was given in

the installation of the anemometers at both towers.

(a) Vertical Distribution of Mean Wind Speed

The data obtained at the Loran Tower during Typhoon No. 7017 shows that the power

index is approximately equal to 0.12 although the wind was not very strong. Similar

1-5
power index values were obtained from the other winds. The wind profile at OHK TV-Tower

is not represented by a single straight line in a logarithmic chart and during certain

runs, the power index is 0.47 in the upper part of the profile and 0.22 in the lower region.

(b) Intensity of Turbulence

Examples of the intensity of the turbulence obtained at the OHK TV-Tower and

the Tower Loran are shown in Fig. 5. The turbulence intensity decreases with increasing

height and also mean wind speed at the OHK TV-Tower but the data shows a different

tendency at the Loran Tower.

(c) Gust Factor

The Gust factor u /U (u : maximum wind speed) obtained at both towers is depicted

m m

in Fig. 6. The factor varies between 1.2 and 1.6, and seems to decrease with increasing height and mean wind speed.

Observation at an Array of Towers

Observation at Tarama Island, Okinawa

In order to obtain data on high winds, which are required for the economic design of

towers for large power transmission lines , a number of observing poles were installed at

Tarama Island in Okinawa (Mitsuta, 1974) . The island is elliptic (about 4 km x 5 km) , and

has a flat surface as shown in Fig. 7. Most of the island, ground is level and is around

10 m high above m.s.l. Okinawa, as is well known, is often subjected to typhoons, which is

why Tarama Island has been chosen as an observation site.

The observation site as shown in Fig. 7, is near the southwestern coast. The observation facility consists of an array of 27 observing poles, installed with 39 anemometers and

an instrumentation house. The configuration of the array is shown in Fig. 8. Twenty- five

poles are distributed with equal separation of 30 m on a straight line of 720 m in length.

The height of the poles are 15 m except one which is 50 m in height and is located in the

center of the array. Two 15 m poles are placed 90 m away from the center in the direction

perpendicular to the base line three-cup anemometers and Aerovanetype anemometers are attached to the poles as shown in Fig. 8. The system is started when the monitoring anemometer

output exceeds a preset wind speed. The wind observation was initiated in the middle of

1972 and has continued since that time.

(a) Correlation Coefficient

The wind, whose direction is perpendicular to the base line, is selected and read

out every 1.5 s for 12 min. Examples of auto and cross correlation coefficient for

the same run are shown in Figs. 9 and 10, in which the wind comes from an inland

direction. The dots in Fig. 10 show the correlations of all possible combinations from

the observing points, and the solid lines show the average values. The average integral

scale of the turbulence, in the longitudinal direction, is about 120 m and in the

lateral direction is about 20 m.

(b) Gust Factor

The Gust factor, as a function of both gust duration and its lateral scale at the

height of 15 m was obtained from the data. Figs. 11 and 12 show an example of the two

1-6
dimensional gust factors for the same run. Variations of space averaged gust factors,

with an averaging time, are shown in Fig. 11 and show the power law approximation.

Fig. 12 shows the time averaged gust factors as a function of average time, which also

shows the power law approximation.

Observation and further analysis are now being conducted, and more detailed results

are expected in the near future.

Observation at Satoura, Tokushima Pref .

Five towers were placed along a line parallel to the coastal line, which runs from SSE

to NNE, at Satoura of Naruto City, Tokushima Pref. (Shiotani, 1969), for the same purpose

as described in Section 2.1. The towers were located at 12, 35, 80 and 190 m from the south

end. They are 40 m in height and were installed with Aerovane-type anemometers. The observations were initiated in 1966, with readings taken every 1 s for 10-13 min.

(a) Correlation Coefficient and Scale

The cross correlation coefficient of the longitudinal wind at the points separated

by ri in the lateral direction, is denoted as R (n, T) , where T is time lag. The data

uu shown as open circles in Fig. 13, is the parameters R^^(^) ' which is defined as R (H/ 0) . In the lateral direction the integral scale of the turbulence is about 50 m

uu representing the wind from the sea, and is about 65 m for the wind from the land. The

integral scale is somewhat larger for the wind from the land than from the sea, at a height of 40 m.

The auto correlation coefficient, R (£) where E is the downwind distance in the uu longitudinal direction, can be estimated from the auto correlation coefficient by use

of Taylor's hypothesis. The results are R iO = R (T) where ^ = UT, where the parauu uu meters R (^) is shown in Fig. 13. The longitudinal scale of the turbulence can then

uu

be estimated using this data. The scale thus obtained using a constant height is 204 m for the wind from the sea and 154 m for the wind from the land. The turbulence becomes greater for the wind from the sea, which is opposite to the result in the lateral

direction. The scale of the turbulence in the longitudinal direction is 3 to 4 times

as large as that in the lateral direction.

(b) Space Correlation Coefficient

Using Taylor's hypothesis, the spatial pattern of the space correlation coefficient

R^^(£i, r\) can be estimated. An example of the pattern is shown in Fig. 14.

Field Observation by Other Methods

In connection with air pollution, as mentioned previously, extensive observations have

been made in industrial and urban areas. In addition to making observations using towers,

other techniques have been employed.

One such method utilizes a special radiosonde attached to a tethered balloon and can measure turbulent winds (Yokoyama, 1969) . A low altitude radiosonde has recently been

developed for observing the wind, temperature and pressure in the layer up to 2000 m.

These instruments and the low altitude radiosonde are often used for observing the wind

over the urban area (Takeuchi, 1975) . Data thus obtained are of great use in classifying

1-7
local wind.

Additional papers concerning the turbulent wind over the urban area have been published,

by Nakano at al., 1974 and Hanafusa et al., 1974.

Wind Tunnel Experiment

In order to clarify the characteristics of local wind, many tests have been conducted

in a wind tunnel, but few papers have been published.

Kamei et al. (1974) conducted wind tunnels experiments in order to study local wind

induced by tall buildings, and also recommended some practical methods to estimate the high

wind region.

Suda et al. (1974) , presented a paper on model experiments concerned with the study of

local winds induced by a small island. This study will be a great help in the design of

long span suspension bridges.

In order to simulate the atmospheric boundary layer in the laboratory, Sato et al. (1974)

developed a velocity-distribution generator by introducing a variable resistance in the wind

tunnel. This device consists of a series of sliding plates, and should be of great use in

studying the local wind in detail.

Concluding Remarks

A survey of recent work on the characteristics and structure of the wind in Japan is

given relative to the author's interests. Thus, other important studies are not included

herein. This survey indicated that further research studies on the structure of turbulent

wind, should be conducted.

Data should be accumulated by various methods (e.g., tower, tethered balloon and any

radiosonde) at different locations i.e. urban areas and coastal areas. The structure of

local wind should also be clarified in relation to the synoptic meteorology and topography.

1-8
References

1. Hanafusa, T. and S. Soma, 1974: On the Characteristics of Atmospheric Turbulence Above

the Urban Area. Proc. of the Third Symposium on Wind Effects on Structures in Japan,

63-70. (in Japanese with English abstract).

2. Kamei, I. and E. Maruta, 1974: The Wind Tunnel Test on High Wind Region Induced With

Tall Building. Proc. of the Third Symposium on Wind Effects on Structures in Japan,

115-122. (in Japanese with English abstract).

3. Kinase, T. , N. Toyama c S. Ito, 1972: On the Structure of Strong Wind for Constructing 600 m Tower. Meteorological Material for Tower Construction (Japan Meteorological

Agency), No. 4, 22-43. (in Japanese).

4. Mitsuta, Y. , 1966: Sonic Anemometer-Thermometer for General Use. J. Meteor. Soc.

Japan, 44, 12-24.

5. Mitsuta, Y., 1974: Preliminary Results of Typhoon Wind Observation at Tarama Island,

Okinawa. Paper presented at US-Japan Research Seminar on Wind Effects on Structures,

Kyoto, Sept. 9-13, 1974.

6. Nakano, M. , K. Takeuchi, Y. Mitsuta and T. Hanafusa, 1974: On the Characteristics of

Atmospheric Turbulence Above the Urban Area. Meteor. Research Note, No. 119, 133-141

(in Japanese)

.

7. Sato, H. , Y. Onda and T. Saito, 1974: Laboratory Simulation of Atmospheric Turbulence:

Generation of Arbitrary Velocity Distribution and Model Experiment of Flow Around Mt.

Fuji. Advances in Geophysics, 18B , 241 - 251.

8. Shiotanti, M. , 1969: Structures of Gusts in High Winds (Interim report. Part 3).

Nihon Univ. at Narashino. 118 pp. (in Japanese with English abstract)

.

9. Shiotani, M. , 1975: Turbulence Measurements at the Sea Coast During High Winds. J.

Meteor. Soc. Japan, 53^. (to be published)

10. Soma, S., 1964: Turbulent Structures in Strong Winds. J. Meteor. Soc. Japan, 42 , 372 - 396. (in Japanese with English abstract)

.

11. Suda, K. , S. Soma and K. Takeuchi, 1974: Wind Tunnel Experiments for Studying a Local

Wind. Paper presented at the Sixth Joint Meeting of US-Japan Pannel on the Wind and

Seismic Effects, Washington, D.C., May 15-17, 1974.

12. Takeuchi, K. , 1975: Observation of Local Wind. J. Appl. Phys. Japan, (to be published)

13. Yokoyama, O. , 1969: Measurement of Wind Fluctuations by a Vane Mounted on the Captive

Balloon Cable. J. Meteor. Soc. Japan, 47^, 159-166.

14. Yodoyama, 0., 1971: An Experimental Study on the Structure of Turbulence in the Lowest

500 m of the Atmosphere and Diffusion in it. Report of National Research Institute for

Pollution and Resources. No. 2.

15. Yokoyama, O. , 1972: The structure of the Planetary Boundary Layer Obtained by

Measurements, Tenki (meteor. Soc. Japan) , 19^, 654-659. (in Japanese)

.

1-9.
Fig.l Topography around the meteorological tower by the mouth of

River Naka (shiotani ,1975)

cr-LO
Ri Zi.Zi ; x) RUN 63

80.8 — sae m

50.8 — 30.8 m

80. 8 ~ 30.8 n

ig.2 Cross correlation coefficients between two heights

{Shiotani,1975)

a44» can 0S74 i3S32 0£50 0591 0507 0410 0329 0259 0203

RUN 47 WIND VELOCITY

0432 aSOE Q53S Q6«7 0692 0630 0547 0464 0422 0373 0.2S0 100 IB

Fig. 3 Isopleth of correlation coefficients in a vertical

plane parallel to mean wind velocity (Shiotani ,1975)
1-12
03

02

0.1

CO

on

1 Ol ltd

T7002/0KINAW

1

A

03

CD

.0.1

OD

A

•

HEIGHT(m)

• 165 • 1345

103 ' 72

o 435* 175

^ va < f

«

f I * U • 4

•

•

•

• •

T700i /OKINAWA

i 50 100 150 200 "^-^16 20 24 28 32 36

HEIGHT (m) (o) MEAN SPEED (m/s)

50 100 150 200 250 300 350 400

HEIGHT (m)

(b)

Fig. 5 Intensity of turbulence: (a) Okinawa, (b) Iwo-Jima

(Kinase et al.,1972)

50 100 150

HEIGHT (m)

T7002/0KI.

HEIGHT Ur\)

. • 165

• 1 34.5

03

I!.5

17.5

16

(a)

20 24 28 32

MEAN SPEED (m/s)

36

50 ICQ 150 200 250 300

HEIGHT (m)

(b)

350 400

Fig. 6 Gust factor: (a) Okinawa, (b) Iwo-Jima (Kinase et al.,1972)

1-13
I

Fig. 7 Topography of Tarama Island, A: observation site

(Mitsuta,1974)

Fig. 8 Arrangement of observing poles and anemometers

(Mitsuta,1974)

1-14
I I t I I 0 IS 30 <<5 60

Time (sec)

Fig. 9 Auto-correlation coefficient of wind velocity whose direction

is perpendicular to the base line (Mitsuta,1974)

Fig. 10 Same as Fig. 9 but space correlation coefficient

1-15
( sec

)

Averaging Time

Fig.n Space averaged gust factors as a function of

averaging time (Mitsuta,1974)

Run 1

No. of ave.

10

Averaging Distance

Fig. 12 Time averaged gust factors as a function of

averaging distance in lateral direction

(Mitsuta,1974)

1-16
ESE (40m) RUN 125

26

KM

Fig. 13 Averaged cross-correlation coefficients as a function of lateral distance (open circle) and

averaged auto-correlation coefficients as a function of downwind distance (Shiotani ,1969)

RUN 2062

WIND

N

14 Spatial pattern of correlation coefficients (Shiotani ,1969)

1-17
A REEXAMINATION OF HURRICANE CAMILLE

by

Arnold R. Hull

Environmental Data Service

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

ABSTRACT

Newly available oceanographic and meteorological data on major storms and hurricanes

striking the Gulf of Mexico during a 31-month period provided by eight petroleum firms present an opportunity to reexamine and reevaluate Hurricane Camille, one of the most severe

and destructive storms ever to strike the Gulf. (Maximum wave heights of 72 feet were

recorded as the eye of the hurricane passed within 15 miles of one measurement station.)

The new data on Camille comprise one of the most comprehensive sets of oceanographic and

meteorogolical information available for such an extreme weather event and should prove

invaluable in basic research and offshore engineering applications. With increasing

availability of this type of information likely as the Nation develops its offshore energy

resources, questions arise as to what procedures should be followed to access, disseminate,

and use these data most effectively and what contributions this important new data source may make to current knowledge of extreme storm events and their effects on engineering

structures. The data available for Camille are examined to provide tentative answers to

these questions.

Key Words: Hurricane; Storm Surge; Tropical Storms; Wind; Wind Data; Wind Speed.

1-18
Introduction

The Shell Oil Company of Houston, Texas, acting on behalf of eight petroleum firms,

recently donated oceanographic and meteorological data on major storms and hurricanes

striking the Gulf of Mexico from October 1968 to November 1971 to the Environmental Data

Service of the U.S. Department of Commerce's National Oceanic and Atmospheric Administration. These data include information on Hurricane Camille, one of the most intense and

destructive tropical storms ever to hit the United States mainland. Camille killed 256

people and caused property damage of about $1.42 billion along the Mississippi/Louisiana

Gulf Coast and in Virginia (flooding). [2]*

The data were collected at a cost exceeding one million dollars in a cooperative

effort called the Oceanographic Data Gathering Program.

Data on wave height, wind speed and direction, and barometric pressure were recorded

at six offshore drilling and production platforms spaced along 260 miles of the Louisiana

coastline. These six locations are shown in figure 1.

Wind and Wave Instrumentation

The wind and wave sensors used were Baylor Model 9737 Wave Measuring Systems (rugged

construction, non-fouling, with a long record of reliable performance) and Bendix Model

120 Aerovanes, which measure wind speed through the rotation of a propeller which drives

a DC generator. The wave-measuring instruments give a continuous analog output, with a resolution of 0.1% of full-scale accuracy and a linearity of 1.0% of full scale. The wind

speed threshold is 3 mph, with an accuracy of +. 3/4 mph from 3 to 45 mph, and +_ 3 mph from

45 to 300 mph. The direction sensors are servo- transmitters driving servo-receivers

which rotate potentiometers. The overall accuracy of wind direction is j;^ 5°, with a 10°

deadband at north. [4]

Data Collected

The total data collection comprises 252 analog magnetic tapes. Additionally, 170

strip charts of varying quality and usefulness were recovered. When estimated by sensor,

data were gathered for waves approximately 70% of the time; for wind speed 61% of the time;

and for wind direction 55% of the time. In general, the quality of the data recorded on the

magnetic tapes is good. The accuracy of the data gathered by each sensor must be considered individually, but the accuracy for all data is generally within j;^ 5% of the value.

Data Previously Available on Camille

In severe storm situations it is difficult to obtain reliable wind speed measurements,

and Hurricane Camille was no exception. Of data previously available, approximately six

wind speed observation records could be considered reasonably reliable. Two of the most

useful in establishing the general level of the wind were the observations taken at

Keesler Air Force Base and a record from Transworld Drilling Company's Rig 50.

* Figures in brackets refer to the references listed at the end of this report.

1-19
A reproduction of the Transworld record [1] is given in figure 2. It shows a maximum

wind speed of 172 mph at the 100 ft. level. When reduced to the 30-ft. level using the

1/7 power law, the resulting peak gust is 144 mph.

Earlier wave data from the storm area are limited to a total of 14 observations from

ships close enough to Camille to record winds of 35 knots or more. [5]

The most complete information previously available on Hurricane Camille concerns storm

surge and is based on charts prepared by the U.S. Geological Survey (figure 3).

New Camille Data

Camille was spawned by a tropical wave that moved off the African coast on August 5

,

1969. It was tracked across the Atlantic to the western tip of Cuba by the evening of

August 15, entered the Gulf of Mexico with a central pressure of 908 millibars, and early on August 17 was located 250 miles south of Mobile, Alabama.

Camille' s storm track (figure 1) was such that the storm passed between Stations 1 and

2 of the Ocean Data Gathering Program, with the center passing about 14 miles west-southwest

of Station 1 at 1730 CDT on August 17, 1969, and about 48 miles east-northeast of Station

2 at 1645 CDT on August 17. [3] Station 1 gathered complete wave data until 1600 CDT on August 17, when the wave sensor broke loose subsea, twisted, and shorted out against the

platform. Wind data continued to be gathered until 1620 CDT August 17 when salt water

shorted out the power system. Subsequent to the loss of power to the station, the Bendix

Aerovane came apart leaving only the center section assembly attached to the tower.

Twenty foot waves were first recorded at Station 1 at about 0500 CDT August 17, 1969,

marking the beginning of the storm at that station. Due to power failure, which occurred

at 1620 CDT August 17, 1969, the end of the storm at Station 1 could not be determined.

Significant wave and wind characteristics during the period from 1330 to 2230 CDT on August 17 at platforms 1 and 2 appear in Table 1.

Wave data have been analyzed for one-half hour intervals during periods of the storm,

and the wave spectrum for Station 1 for the period 1545 to 1615 CDT August 17, 1969, appears

as figure 4. At that time, Camille was located about 24 miles south-southwest of the

station, and station weather conditions are thought to have been very close to their

maximum values. The significant wave height was calculated from each wave spectra and

figure 5 shows the significant wave heights versus time for Stations 1 through 3. The

maximxim wave height recorded during the same interval (1545 to 1615 CDT) was 72 ft.

Significance of the New Data

Using the extreme value distributions of wind speed and significant wave heights for

extratropical storms available from Ocean Station Vessel data, H.C.S. Thom [6] developed

a technique to derive an extreme wave height distribution in any extratropical area of the

deep oceans. With new data such as that for Hurricane Camille, where the relationship of

wave scale to wind scale can be determined, a similar technique might be developed for

tropical areas. And such data are likely to become increasingly available as the United

States and other nations expand their offshore energy exploration and drilling activities.

1-20
Recognizing the importance of wind, wave, and storm surge information to structural

design along coastal areas and to offshore facilities such as nuclear powerplants, drilling

rigs, and supertanker terminals, it is imperative that immediate attention be given to

assessing specific requirements for such information in our coastal zone areas. The increasing number of offshore platforms that will soon be erected in coastal waters offers

us the opportunity to develop onsite observation networks capable of obtaining the optimum

wind and wave information needed to give us a better understanding of the internal structure of hurricanes and tropical storms. Such onsite data collection will provide at least

semipermanent sources of detailed "point" observations, as opposed to current random ship

observations summarized over one degree squares (or even larger areas) because of the lack

of fixed observation platforms.

Such long-term "point" records are critical to the development of accurate detailed

environmental data statistics needed for the design, construction, and operation of both

offshore and onshore facilities in the coastal area.

1-21
References

1. DeAngelis, R. M. and E. R. Nelson. Hurricane Camille-August 5-22, Climatological Data,

National Summary , Vol. 20, No. 8, ESSA, Washington, 1969.

2. Dikkers, R. D. , R. D. Marshall, and H. C. S. Thorn. Hurricane Camille-August 1969,

NBS Technical Note 569 , NBS Washington, 1971.

3. Hamilton, R. C. and D. B. Steere. Ocean Data Gathering Program Report No. 2, Covering

Hurricane Camille, Baylor Company, Houston, 1969.

4. Hamilton, R. C. and E. G. Ward. Ocean Data Gathering Program - Quality and Reduction

of Data, 6th Annual Offshore Technology Conference, Paper No. 2108-A, Houston, 1974.

5. Selected Gale Observations, Mariners Weather Log , Vol. 14, No. 1, Essa, Washington,

1970.

6. Thorn, H. C. S. Extreme Wave Height Distributions over the Oceans, Environmental Data

Service, NOAA, 1972 (an unpublished report).

1-22
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1-24
Figure 2. This wind chart from Transworld Drilling Company's Rig 50 was set on double scale, and the recorder was left running

after the crew evacuated.

heiglits above mean sea level

groy area indicates extent of fluodinK

Figure 3. These are the results of the storm surge generated by Hurricane

Camille on the Gulf Coast from the Pearl River to Bayou La Batre;

based on charts prepared by the U.S. Geological Survey.

1-25
WAVE POl'JER SPECTRAL DENSITY-LINEAR

Data Tape 115

Location SP 62 A

Time 1545-1615 CDT August 17, 1969

CYCLES PER SECOND- HERTZ

Area under curve, Mq = 116. 82 ft. ^ R.M.S. wave height, Vl^ = 10.8 ft. Significant wave height, 4\!Mq = 43.2 ft. Period of Max. Energy = 13.2 sec. Figure 4

1-26
SIGNIFICANT WAVE HEIGHT VS. TIME

Figure 5

1-27
FIRE TORNADO AND ITS MAXIMUM WIND-SPEED

by

S . Soma

Physical Meteorology Laboratory

Meteorological Research Institute

and

K. Suda

Director

Meteorological Research Institute

ABSTRACT

Very few studies have been conducted on fire tornadoes, owing to its rare occurrence.

However, in Japan, much public attention has been directed to this phenomenon in view of the

fact that catastrophic damage has been caused by two fire tornadoes. One such tornado

developed during the Great Kanto Earthquake in 1923 and another at Wakayama City in 1945.

The cause of such catastrophic damage, due to these fire tornadoes is two- fold. One is the

formation of a tornado, in burning of the urban area, and the other is the peculiar feature

that the tornado is accompanied by strong winds.

In the present report, the characteristics of the fire tornadoes, which have been

experienced in Japan are reported and the estimated maximum wind speed in the tornadoes are given

.

Key Words: Fires; Fire Tornados; Tornados; Tornado Model; Wind Speed.

1-28
Introduction

Only a very few studies have been made on the fire tornadoes because of the rarity of

the phenomenon and accordingly, many problems related to it are left unsolved. However, in

Japan, much public attention has been directed to this phenomenon in view of the catastrophic events that occurred during the Great Kanto Earthquake in 1923, which killed 38,000

people. The formation of this fire tornado was then attributed to the passage of a cold

front and this view has been maintained for about fifty years without any reexamination by

scientists. However, a survey of studies made in various countries and the detailed analysis of weather data at the time of the formation of the tornado, revealed that the fire

tornado was due mainly to the conflagration caused by the earthquake.

In Japan, two well known fire tornadoes have caused many casualties: one is related to

the Great Kanto Earthquake in 1923 and the other to the great fire of Wakayama in 1945.

The causes of the catastrophic damage are: (1) formation of the tornado in the burning of

the urban area and (2) peculiar feature of the tornado accompanied by extremely strong wind.

In the present report the formation and characteristics of the above two tornadoes, in

Japan, are described.

1. Process Leading to the Formation of Fire Tornado in Tokyo.

(i) The Old Site for the Clothing Depot just before the Formation of the Fire Tornado

In the City of Tokyo, in 1923, there was a vacant area of 70,000 square meters

wide along the River Sumida, which had once been the site for the Clothing Depot of the

Imperial Army and then transferred to the Tokyo City for construction of a park. This

vacant area popularly called "The Old Site for the Clothing Depot" (to be abbreviated

OSCD) was located in downtown Tokyo which was crowded with wooden houses. Following

the great shock of September 1, 1923, fire spread all over this area and refugees

from the fire advanced toward the vacant ground of the OSCD voluntarily or led by

policemen. As a result, a little after 3 p.m. on Sept. 1, the ground was crowded

with more than 40,000 refugees. People there, free from fear, were eating food or making a bed for the night and nobody expected the formation of a fire tornado.

(ii) Spread of Fire Around the OSCD.

Examination of the spread of the fire, on the basis of a map of the burnt area prepared by Prof. S. Nakamura of Tokyo University, reveals that about 3:30 to 4;00 p.m. on Sept. 1, 1923, when the fire tornado formed, the entire Tokyo City had not been on

fire. Around the OSCD along the River Sumida, however, a vast urban area of 3,000

meters long in North-South and 1,600 meters wide in East-WEst had already been on

fire except the OSCD. In other words, the OSCD was then surrounded by a burning area

several times larger than itself and was crowded with 40 thousand refugees.

(iii) Strike of the Fire Tornado

A refugee who barely survived the disaster of OSCD and witnessed the strike of

fire tornado reports the following: The tornado attacked OSCD with a deafening roar

and scattered fire-flakes on the heads of the refugees. Clothes and household goods,

brought by the refugees caught fire at once and the fire, fanned by strong wind,

spread all over the OSCD. The tornado raged for only about 20 minutes but it took

1-29
38,000 lives in a quick moment.

2. Characteristic Feature of the Fire Tornado

(i) Shape of the Fire Tornado

Based on old literature, a study was made on the shape of the OSCD fire tornado

in order to see whether it was like the funnel of a natural tornado hanging down from

the mother cloud.

There were at least two thousand people who barely survived the rage of the

strong wind and flame in OSCD. They must have had a chance to witness the fire tornado by their own eyes. Furthermore, there must be many people who could see the

tornado from the opposite bank of River Siimida, even if they were frightened by the

earthquake. As the river is only 150 meters wide, it must well be possible to observe

the tornado across the river. But, only very few of the old literature includes a description of the shape of fire tornado, as quoted below.

"About 3:30 p.m. I saw the smoke gathering and towering high up in the sky"

witness of Mr. K. Kondo. "About 3:30 p.m. a strong crack was heard from the northern

direction. With the intensification of the sound, some attributed it to approaching

tsunami while others to ignition of explosives but nothing definite was known. Then

I saw deep black smoke flowing in a whirl as big as a hill presenting a horrible

sight." —witness of Mr. K. Kawanabe who observed the fire tornado from the Ekoin ground

located 700 m to the south of OSCD.

These witnesses may perhaps suggest the presence of a whirling fire tornado but

nothing definite is known about its shape.

(ii) Wind Speed Associated With the Fire Tornado

The order of magnitude of the wind speed, associated with the fire tornado, is

a matter of meteorological interest but anemonetric observation is never possible in

view of the condition of the tornado formation. In the case of the OSCD fire tornado,

some refugees testified that persons and carts were picked up by the strong wind, but

it is difficult to estimate the wind speed from the testimony alone. Fortunately,

photographic plates showing the wind damage on trees were found in the Report of the

Great Kanto Earthquake and we could estimate the maximum wind speed associated with

the fire tornado. Plate 1 (a) , which is reproduced from the above report, shows that

the trunk of a tree, about 30 cm in diameter, was broken by the wind. A picture of

the wind damage of trees, quite similar to the above, was taken during the tornado

which formed in July, 1971 in the suburbs of Omiya City, Saitama Prefecture (Plate 1

(b) ) . According to K. Watanabe of the Meteorological Research Institute (1971) , the

intensity of this tornado is estimated to rank among Class F3 of Fujita Scale, which

ranges from 70.4 to 92.5 m/sec. Considering the similarity of wind damage of trees

in both cases and referring to the statements of witnesses quoted above, we may suppose

that the fire tornado of OSCD was accompanied by strong winds of the same order of

magnitude

.

1-30
(iii) Direction of the movement of the OSCD fire tornado

Examining the reports on the Great Kanto Earthquake , we found that unexpectedly

many people stated the direction of the fire tornado. However, many have mistaken the

direction of the wind, for the direction of the tornado motion. Dr. S. Fujiwara (1924)

made a comparative study of their reports and estimated the track of the tornado wind

system as follows

:

"Judging from the above circumstances, the fire tornado that brought about the

disaster in OSCD would have formed at the southern border of Kitabanbacho , passed nearby

Umayabashi Bridge, moved along the River Sumida up to Mr. Yasuda's mansion in Yokoamichcy

turned eastward, merged with other tornadoes at Ishiharcho and then moved back to

OSCD from the northeast." If we trace the motion of the tornado based on Dr.

Fujiwhara's analysis, the tornado seems to have advanced from northeast to southeast.

No natural tornado ever observed in Japan moved in such a direction. Furthermore, the

direction of motion of the OSCD tornado was exactly opposite to the direction of the

prevailing wind.

(iv) Clouds in the Case of the Great Earthquake

In the old data log of the Central Meteorological Observatory in Tokyo, Cb

cloud of 3/10 in amount, is recorded for 4 p.m. of September 1. It is almost certain

that this cloud was then overhead of the OSCD. The statement in the Report of Great

Kanto Earthquake "As it began to rain, we made preparation to protect us from it with

sliding doors and straw mats." supports this fact. The cloud, which is recorded in the

data log of the Observatory simply as Cb, seems to have been unusual enough in shape

to attract public attention and is mentioned in many statements in the Report of the

Earthquake. From these. Prof. T. Terada's statement depicting the cloud condition is

quoted below:

"About 3:30 p.m. on September 1, I stood in the verandah of my house at

Komagome-Akebonocho , looking at a remarkable cumulus cloud towering up in the sky in

the southeast. Surface appearance of the cloud was characteristically different from

the average, i.e., protuberances on the surface were fine and well-defined reminding

me of the texture of the volcanic smoke photographed during the recent eruption of

Mt, Sakurajima. The towering of the cumulus cloud in the blue sky presented an un- precedented spectacle,"

This cumulus cloud may be identified with the cloud that covered the OSCD since it

was observed by Prof. Terada from Kcmagome Akebonocho in the southeastern direction.

A picture of cumulonimbus taken by Mitso Harada was found in the Report of the

Earthquake (Plate 2) . It is not clear whether this was the cloud covering OSCD, but

its unusual shape is exactly the same as depicted by Prof. Terada. Although the

formation of the cloud may partly be due to the favorable synoptic condition such as

the presence of subtropical air mass and front, its direct cause will more properly be

attributed to the energy of instability created in the stratification of the atmosphere

by the extensive fire.

1-31
3. Examples of Fire Tornadoes

In the history of disasters in foreign countries violent fire tornodoes, that have

burnt to death tens of thousand people have been recorded. However, smaller ones have been

recorded, and formed in olden times as well as in recent years. One such example of a foreign fire tornado is given herein;

In June, 1961, a French cloud physicist J. Dessens conducted an experiment by making

artificial cumulus clouds on the plateau of the Pyrenees. He placed fourty-eight kerosene

burners in an area of 125 x 125 km wide and make lit all of them at once. Then, quite un- expectedly, a fire tornado formed in the lee of the experiment area and developed to a size

of 10 meters in diameter and 200 meters in height in a distance of 525 meters leeward from

the origin of the fire (Plate 3- (a)). Careful observations revealed that, following the

first tornado, a number of others formed in succession.

In May, 1965, in the port of Muroran, a Norwegian mammoth tanker (56,000 t) cracked in

a wharf. A large hole was then made on the side of the ship and petroleum gushed out and

caught fire. In this case too, a remarkable fire tornado formed in the lee of the large

fire. (Plate 4)

On the occasion of the Nankai Earthquake (magnitude 8.1) in December, 1946, a large

fire broke out in Shingu City, Wakayama Prefecture, and burnt down 2,612 houses, destroying

25% of the entire urban area. On this occasion, a fire tornado is reported to have formed

in the river beach, in which citizens took refuge. This tornado was not large, but was accompanied by a strong wind which blew up a mat of 22 Kg in weight into the sky.

4. Fire Tornado in the Great Fire of Wakayama City

(i) Formation of the Fire Tornado

A fire tornado, much larger than those mentioned above, formed in association

with the great fire of Wakayama City in July, 1945, which destroyed 68% of the urban

area. On this occasion, a violent fire tornado formed in vacant area of the old prefectural offices, and burnt to d-eath 748 people. Although the number- of victims was

far less than that of the catastrophic OSCD tornado, it corresponds to more than 60%

of the total loss of lives due to the fire, showing the violence of fire tornado.

Now, several statements on the formation of the tornado in Wakayama are quoted

from old reports . "Thirty to fourty minutes after the outbreak of the great fire,

violent fire tornadoes formed in several places in the city. The largest one among

them originated in a street with a tramway 200m away to the south of the old city offic

ground and moved northward. The incandescent flame of the fire tornado streamed

horizontally as if it were blown by bellows. Pine trees planted along the moat were

torn away. An automobile flew over the bank and jumped into the moat; red hot zinc

sheets and big logs were uplifted to a height of 20 - 30 meters."

Aside from the above, another fire tornado formed at night on the same day over

a spot, 1500 meters away from the old prefectural office ground. "In the Naval timber

yard there was hundreds of thousands of feet of timber stored. This timber caught fire

and were uplifted by the tornado presenting a spectable of countless burning chopsticks

flying high in to the sky to a great astonishment of the refugees around Tsukiji Bridge

1-32
"According to the observations, taken by Wakayama Weather Station, east-northeasterly

winds of 2.5 m/s was blowing just before the outbreak of the fire. Examining the

weather charts, six hours prior to the fire, we find that fine weather prevailed all

around Wakayama City without any indication of front or other weather systems , that

would lead to the formation of a tornado,

(ii) Estimation of Maximum Wind Speed

It is not easy to estimate exactly what the maximum wind speed was in association

with the fire tornado of Wakayama City. However, let us try to make a rough estimate

using recent studies on Tornadoes.

To begin with, the estimation of the damage of the pine trees planted along the

avenue, wind speed of about 40 m/s were required to break the pine trees. However,

from the expression that "pine trees—were broken" and that "big logs were lifted up to

the height of 20-30 meters", the maximum wind speed would probably have exceeded 50

m/s

.

Next, turning to the wind estimation by examining the damage of automobiles,

we refer to the record of the tornado that struck Toyohashi City, Aichi Prefecture in

December, 1969. As this tornado passed right across damage, not only to the buildings

of the drive-in and its annex but also to the cars parked therein. Some of the cars were reported to have blown a distance of some tens of meters. Ishizaki and others

(1970) made an estimation of the maximum wind speed accompanying the tornado based on

the damage caused on buildings and concluded that the speed was in the range from 63

to 119 m/s (Fig. 1) . If the estimation is applicable to the damage caused on cars by

the fire tornado of Wakayama, it will be possible to rank the intensity of the tornado

on a scale of F2, which corresponds to wind speeds ranging from 50.4 to 70.3 m/s.

For the wind speed of the fire tornado, which lifted up numerous timbers at the

Naval timber yard in Wakayama, we shall only make an estimate of the speed of the updraft. If we suppose that the diameter of the timber lifted up by the wind to be 20 cm,

3 density of the timber to be 0.5 gr/cm and the drag coefficient (C^) of a cylinder to

be 1.0, then we get 35 m/s as the speed of the upward current that will hold the

timber from falling down. As for the horizontal wind speed, we have at present no

means to estimate its value.

5. Experimental Study

Although a considerable number of fire tornadoes have been observed, the mechanism of

their formation has not been made. It is difficult to clarify the mechanism through the

study of weather data and other materials in the formative stage of the tornado because of

the peculiarity of the condition during the formation of the tornado. Therefore, an experimental study of fire tornadoes was made. These experiments, which are incomplete and have

yielded only qualitative results, will now be outlined,

(i) Experimental Apparatus

The apparatus used in the experiment consisted of a circular metal base, 2 meters in

diameter simulating the ground surface and a wind generator that can provide any wind

distribution in either the horizontal and vertical directions (Plate 5) . Attached to

1-33
bottom of the metal base, is a tank of 2 meters in diameter which was filled with water.

The water temperature was controlled in order to give instability to the stratification

of the air above the base. The wind generator has 15 wind ducts arranged in three rows and five lines, each of which is equipped with a fan and can produce any desired wind

speed.

(a) The formation of a fire tornado is developed by placing on the metal base a circular burning plate, which simulates in the simplest way the burning urban area.

The burning disc, 600 mm in diameter and 10 mm thick, is made from glass-wool

soaked with methyl alcohol.

Raising the temperature of the surface of the base by 20 °C above the room temperature and, at the same time, giving an appropriate wind speed, a fire tornado

of Dessens type was formed as shown in Plate 3 (b)

.

(b) Considering the fact that, in the formation of the OSCD tornado, the OSCD

ground was surrounded by a burning urban area, an experiment was conducted using

a burning plate similar to the burning area in shape. In the same manner as described previously, the temperature of the surface of the metal base was raised

by 20°C above the room temperature. Then, after giving appropriate wind speed and

horizontal wind shear, formation of a remarkable vortex column like a whirl-wind

was observed at the spot surrounded the burning area (Plate 5 and Fig. 2)

.

Although this experiment simplifies the natural wind condition, it seems to give

essentially the model of the osCD fire tornado.

If the temperature of the metal base is lowered to decrease the instability of

air stratification, the formation of the vortex column like tornado becomes rare.

However, even if the metal base is not heated, vortex columns are formed sporadically at the same spot, after applying an appropriate wind speed and wind shear.

5. Characteristic Feature of the Earthquake Fire

Our investigation of the fire tornadoes in the OSCD and the Wakayama prefectural

office ground revealed that the presence of an extensive burning area seems to be the

decisive factor in the formation of fire tornadoes. However, not all of the large fires

recorded in the history are accompanied by fire tornadoes. Although further investigation is required for the final solution of this problem, it is highly probable that there

are at least two kinds of fire tornadoes, one favorable for the formation of a fire and

the other not.

Examining materials of past large fires, it was found that in most cases strong wind

played an important part in the extensive spread of the fire. For example, in the instance

of the great fire of Hakodate City, Hokkaido, which destroyed the major part of the city on

March 21, 1934, the maximum wind speed of 22 m/s was observed. Among the Three Great

Fires of Edo (old Tokyo) , which are well known in the modern history of Japan, at least two

were due to strong wind. A common feature of these is that the fire originated from one

source and spread to an extensive area fanned by a strong wind; the burned area extends

in a strip (Fig. 3) . However, fires due to earthquake show a burn as an area with quite a

1-34
different shape. In this case, there are numerous fire sources and the fire breaks out all

at once over an extensive area. In the case of the great fire of Wakayama, which was not

due to an earthquake, the shape of burned area indicates the same process. Therefore, it

can be concluded that there are two kinds of great fires, one of which fire breaks out

from one source and fanned by strong wind, and then extends along a strip. The other

fire starts burning all at once from numerous sources. Generally in the formation of fire

tornadoes, the latter is more favorable.

In describing the experimental study on fire tornadoes, we have stated that a vortex

column like fire tornado was formed providing an appropriate wind speed was given. Namely,

in our experiment, the vortex column did not form in a weak wind or a strong wind, but did

form if a wind speed around 1 m/s was provided. As a matter of fact, both of the fire

tornadoes of OSCD and of Wakayama prefectural office ground formed in relatively weak wind

with a speed of 4 - 5 m/s for the former and 2-3 m/s for the latter. Thus the wind speeds

of such range seem to be appropriate for the formation of fire tornadoes.

Conclusions

Fire tornadoes of OSCD and Wakayama prefectural office ground, which caused catastrophies

were studies relative to their cause formation, their characteristic features and their

relation to the damage induced. As a result, it was found that the main cause of their

formation was the extensive fire that was initiated from numerous sources, the peculiar

shape of the burning area, an appropriate wind speed and the unstable stratification of

the atmosphere. Furthermore, it was revealed that fire tornadoes are accompanied by very

strong winds reaching speeds of 70.4 - 92.5 m/s for OSCD tornado and 50.4 - 70.3 m/s for

Wakayama prefectural office ground.

The major characteristic of fire tornadoes is the updrift and induced fire-flakes, thus

spreading the burnt area. In addition, the fire tornado fans the fire with violent wind

that surpasses the storm in typhoon.

In the report of the OSCD fire tornado, the following statements are found. "Fire

creeped about over the ground. People fell down breathing in the flame." "He survived

the fire without getting burnt but, a few days later, died complaining of a pain in his

breast". There are a few other statements with the same implication. According to "The

Fire" compiled by Dr. K. Nakata (1969) , the lung may be burnt by breathing in fire which

can be fatal to life. At first difficult breathing occurs due to the break of pulmonary

capillaries, then emphysema of lungs. Patients, if they survive this stage, they are generally doomed to death suffering from bacterial pneumonia.

In the case of the OSCD fire tornado, an exceedingly strong wind in the open area

made the fales creep about the groiind and consequently many people would probably have

breathed in hot air. Recently, if the fire of tall buildings, many people have been

reported to have been suffocated to death by breathing in smoke. In the case of fire

tornadoes, however, the process leading to death is somewhat different.

1-35
References

1. Central Meteorological Observatory (1924) : Report of the Great Kanto Earthquake

(Meteorological Part, in Japanese) , pp. 1 - 151.

2. Dessens, J. (1962): Man-made Tornadoes, Nature, Jan. 6, Vol. 193, pp. 13-14.

3. Isizaki, H. et al (1970) : Investigation on Tornado and its damage which struck the

city of Toyohashi, Dec. 7, 1969, WDD Technical Note No. 5, by Grant in-Aid for

Fundamental Scientific Research from the Ministry of Education, (1) -96100.

4. Lee, S.L. (1972): Fire Research, Applied Mechnaics Review, May, Vol. 25, No. 25, pp.

503-509.

5. Nakata, K. (1969): The Fire, Kyoritsu Publishing Co. (in Japanese), pp. 629.

6. Disaster Science Research Society (1956) : Great Fires in Japan, Gihodo Book Co.

(in Japanese) , pp. 13.

7. Terada, T. (1925) : On the tornado of September 1 - 2, 1923, Report of EArthquake

Disaster Prevention Society, No. 100 (in Japanese), pp. 185-227.

8. City Office of Wakayama (1956) : History of War Damage of Wakayama City (in Japanese)

pp. 30-40.

1-36
0 so 100 isom

DISTANCE FROM THE CENTER OF TORNADO

Fig. 1: Maximum wind speed estimated from

the damages of structures (Toyohashi

Tornado December 7, 1969).

(by H. Ishizaki)

Tornado - lika vortex

Wind diraction

Fig. 2:

Relative position of vortex

column and burning areas in

the case when the most remarkable vortex column was generated.

Fig. 3:

An example of the greatest fire

in Tokugawa Era; the fire broke

out at one point and spread leeward in a long and narrow strip

fanned by high wind.

I- 37
Plate 1 (a)

:

A tree torn off by the

fire-tornado. This was

taken at the former site of

the Army Clothing Depot immediately after the Great

Kanto Earthquake.

Plate 1 (b):

A tree torn off by the

meteorological tornado.

This was taken at the

suburbs of Oraiya City, near

Tokyo on 7th July, 1971.

Plate 2: Gigantic cumulonimbus in the sky over Tokyo after the Great

Kanto Earthquake (taken by M. Harada) . The former site of

the Army Clothing Depot seems to have been covered by such

a cumulonimbus at the time of fire-tornado occurrence.
Plate 3 (a) : A fire-tornado which formed

accidentally in the lee of

a large scale fire plume

(by J. Dessens, 1961)

.

Plate 3 (b):

Vortex column like a fire-tornado

in laboratory (by S. Soma, 1973).

1-39
]

Fire-tornado which formed on the leeside

of a burning tanker (56,000 t) in Muroran

Port in Hokkaido, 1965. (Provided by

Hokkaido Newspaper)

Apparatus for the experimental study of

fire-tornado. Vortex column like firetornado is made on the copper disk.

1-40
Plate 6:

Vortex column like fire-tornado

obtained by the burning area

model. Vortex column can be

seen on upper left. (Picture

taken from the above.)

I-4l
A RESEARCH PROJECT ON THE WIND FLOW

AROUND TALL BUILDINGS

by

Tatsuo Murota

Senior Research Officer

3rd Research Division

Building Research Institute

Ministry of Construction

and

Kiyoshi Nakano

Head

Senior Research Officer

3rd Research Division

Building Research Institute

Ministry of Construction

ABSTRACT

The problem of wind effects, produced by the construction of tall buildings, is analyzed. A research project designed to obtain information about this problem and to study

the counterpart for the prevention of high winds around tall buildings, has been initiated.

Some preliminary observations are described herein.

Key Words: Building; Wind; Wind Effects; Wind Loads; Wind Observation; Wind Speed.

1-42
Introduction

During the past 10 years in Japan, the number of tall buildings which have been constructed has increased dramatically. Many of these buildings have been constructed in the

central part of large cities, and sometimes produce high wind speeds near the ground

surface in pedestrian areas. Previously tall buildings were constructed in low building

areas. After construction, alterations have occurred between the owner of the tall building

and nearby habitants about the change of wind environment. It is difficult, however, to

find a solution to the problem, because quantitative data concerning the modification of

air flow and the frequency of occurrence of high winds is too scarce.

The Ministry of Construction, therefore, initiated a research project to obtain

information about this phenomenon and to obtain a guidance on this wind problem.

In the followings, after discussion of the problem concerning this phenomenon, an

outline of this research and some preliminary results of wind observations are described.

Wind Surrounding Tall Buildings

The problems due to the wind caused by the modification of airflow surrounding tall

buildings can be divided into two categories: 1) the change of the environment, and 2) the

increase in damage to the structures surrounding the tall buildings.

The former is concerned with two kinds of phenomena; a) an increase of the frequency

of occurrence of high winds and b) an increase of frequency of occurrence of calm conditions. The "high wind" condition is a wind which has a speed which becomes annoying to

people, which is generally 5 m/s or greater. These "high winds", cause difficulty in

walking, increase heating consumption in winter, create difficulties in opening of doors,

cause disturbance in rooms when windows are open, and impede plant growth. The occurence

of the high winds increase the frequency of calm conditions , and thus , the complaints of

sultriness during the summer.

The change in the environment is the most serious problem, because this is related to

calm wind which occurs very frequently. These winds are affected by the air's thermal

stratification (vertical temperature gradient) in addition to the large-scale weather condition, and thus, makes it very difficult to estimate the change in the environment.

The second wind problem, the increase in damage to structures, is related to higher

winds such as those caused by typhoons. In general, buildings exposed to such high winds

will be subjected to some damage risk. However, buildings standing near tall buildings may

be exposed to the additional increase of wind speed caused by the modification of air flow,

thus increasing the risk of damage.

In order to solve these problems the following information needs to be collected;

Change of Environment

(1) Distribution of wind speed around tall buildings in relation to the air's thermal

stratification.

(2) Frequency of occurrence of each pattern of wind speed distribution.

(3) Criteria of allowable change of environment stipulating the correlation of wind

speed and frequency of occurrence.

1-43
Increase of Damage

(1) Distribution of wind speed around tall buildings

(2) Distribution of turbulence

Description of the Research Project

The Building Research Institute has instituted a research project relative to this

problem since 1975. The purpose of this project is to provide informations stated above

and solutions to eliminate the high wind speeds. The following studies are part of this

project;

(1) Wind observation around tall buildings

High winds around tall buildings have been observed at several tall buildings sites

in Japan. The periods of these observations, however, were for only a few hours, and thus

the data obtained are not reliable in order to understand the quantitative nature of winds

around these tall buildings. Therefore, continual wind observations around and in the

vicinity of a few tall buildings will be conducted during the next several years.

(2) Field investigation of the influence of the wind in the environment

In the vicinity and around many tall buildings, the intensity and spatial distribution

of the change of wind will be studied. The relation between the dimension or shape of tall

buildings and the change of wind on the environment will be analyzed.

(3) Study of wind through wind tunnel test

If a reliable similarity law is available, wind tunnel tests are quite effective in

evaluating the wind distribution. However, such a similarity law is not known at present

and therefore the possibility of such a law will be studied in this project by comparing

ad hoc testing methods with full-scale measurements.

(4) Development of methods to reduce high winds

Windbreaks and shelterbelts have been utilized in agriculture areas in order to reduce

surface wind speed and to obtian the resulting micro-climatic change in plant growth. In

this project such applications will be examined relative to winds on buildings.

Preliminary Results

The Building Research Institute has conducted wind observations in 1974 at two 14-story

building sites.

(1) Wind observations at Tokorozawa City

Tokorozawa New Town is a housing area developed in the outskirt of Tokorozawa City.

Three 14-story apartments are arranged in a line running from west to east in the center of

this New Town. Surrounding the three tall buildings is a five story apartment and one and

two story houses. Wind observations were made at 3 levels of two observation points as shown in Fig. 1. Between these two points a wind-shelter fence of 4 m in height and 40 m

in length were arranged from south to north.

Wind data collected at the Tokorozawa Fire Station, which is located 3 km from the

New Town, were used as the reference location. The height of the anemometer at the

reference point, is 9 m above the ground.

1-44
Fig. 2 shows the frequency of occurrence of wind speeds at points to P in the New

1 6

Town compared with that at the reference point. According to this figure, at P^, P^, P^

and P^ the frequency of occurrence of the calm wind is 10% below 2 m/s. However, the

5

frequency of occurrence of the higher winds is several percents higher at every wind speed.

The frequency of occurrence at points P^ and P^ is quite different from other points and is

similar to that at the Fire Station. This is probably due to the shelter effect of the

fence.

Fig. 3 shows the relationship between the ratio of the wind speeds at points P^ to P^

1 o to the reference wind speeds, when the reference wind direction in N, NNW and NW. The

values of the wind speed ratio R is greatly scattered at the lower reference wind speeds,

but the scatter becomes less with increasing reference wind speeds. The maximum value of R

is greater than 3 and is observed at reference wind speed of 2 m/s. At the higher reference wind speeds the value of R approach unity.

(2) Wind observation at Koto-ku, Tokyo

The other location where preliminary wind observations were conducted was at Koto-ku,

Tokyo. At this location there are eight-fourteen story apartment buildings which were

constructed in 1975 as shown in Fig. 4. Wind observations were made in 1974, when the

apartments No. 4, 5 and 6 were under construction and they were only two storied buildings

at the end of the year. The average wind speeds were continually observed at 6 points as

shown in Fig. 4 and at a height of 5 m above the ground. The wind data collected at a wind tower (54m high above the ground) of the Japan Meteorological Agency, which is located 7 km from the observation site and was used as a reference

.

Fig. 5 shown the relationship between the reference wind speed and the wind speed

ratio R (wind speeds at points P^ to P^ to reference wind speed) , when the reference wind

direction was NW. As shown in this figure, there is a tendency for the R value to scatter

and is similar to that in Tokorozawa areas with the mean value of R differing greatly from

point to point.

1-45
a.d UiO. T

J

B

m.

ELEVATION

PLAN

Rg. 1 ARRANGEMENT OF WIND OBSERVATION POINTS

Pi -P6 . HEIGHTS OF Pi TO Pe ARE 2/,6,2/.5

AND 7m . RESPECTIVELY.

Fig. 2 FREQUENCY OF HOURLY WIND SPEED AT

TOKOROZAWA NEW TOWN IN THE PERIOD

OF DEC.16.1974 TO JAN. 13.1975.

Reference Point (Fire Station)

Pi

^ "J 0~2~2~3 a-A A~5 5-6 6-7 7-8 8-9

WIND SPEED (m/s)

1-46
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WIND SPEED AT REFERENCE POINT (m/s)

Fig. 3 RELATION BETWEEN THE WIND SPEED RAHO R

AND THE WIND SPEED AT REFERENCE POINT

1-47
1-48
00

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1-49

WIND ENGINEERING RESEARCH PROGRAM

SUPPORTED BY

THE NATIONAL SCIENCE FOUNDATION

by

Michael P. Gaus

Head

Engineering Mechanics Section, NSF

Washington, D.C.

ABSTRACT

The National Science Foundation is an Agency of the Federal Government established in

1950 to advance scientific and technical progress in the United States. The Foundation

fulfills this responsibility primarily by sponsoring scientific research, encouraging and

supporting improvements in science education, and fostering scientific information exchange.

NSF does not itself conduct research or carry out education projects.

Operating under this charter the Foundation supports a substantial amount of research

in Wind and Seismic effects. The Division of Engineering has for the past years had an organized area activity in Wind Engineering Research. The primary emphasis of this research

has been toward developing a better understanding of the structure and flow of the earth'

s

boundary layer to develop new knowledge and techniques for predicting and coping with the

interaction between the boundary layer wind and man-made or natural objects and to develop

methods to assess or predict environmental effects related to wind flow. Almost all of the

current research programs being supported are at academic institutions. A summary of these

cvurrent research projects is given herein.

Key Words: Education; Research; Research Programs; Wind Engineering; Wind Studies.

m

ii-i
Introduction

For purposes of classification, Wind Engineering research is divided into 16 categories.

These categories are shown in Fig. 1. Primary emphasis during the early years of the program has been in categories 1 thru 10 and in category 16.

Of special interest during the last year has been the beginning of operations of the

Wind Engineering Research Council which was funded during FY 1974. Objectives and organization of the Council are shown in Figures 2, 3, 4, 5 and 6. The Council has already been

active in promoting interchange between wind engineering researchers and anticipates organizing a series of workshops in the future to assess important problems and opportunities for

wind engineering research.

Also made available during the last year is the first issue of the Wind Engineering

Digest. The Wind Engineering Digest is a survey of current projects in various aspects of

wind engineering research. The digest is intended to alert investigators to research in

progress. The original intention was to include in the first volume only research projects

in the U.S.A. However, because of international interest, it was decided to include international projects for which questionnaires were received. The cover and index of the WERD

Digest are shown in Fig's. 7 and 8.

Activities of the Committee on Natural Disasters of the National Academy of Engineering

have continued to be supported. The wind related activities are twofold. One is to send

investigation teams to sites of natural disasters (such as tornadoes or cyclones) to collect

data on damage and effects which may latter be used to reconstruct damage mechanisms , velocities, missile densities, and so forth. Much of this information is of a perishable

nature and may be obliterated by rescue and clean-up operations. This information is then

made available to researchers who might wish to do in-depth studies on the effects of such

storms. Several teams were dispatched during the last year to document damage due to

tornadoes. The second wind related activity is to prepare a comprehensive report on Wind

Engineering research. This report would be similar to the 1969 report on Earthquake

Engineering Research prepared under the auspices of the same committee which at that time

was operating under the title "Committee on Earthquake Engineering Research." The title

page is shown in Fig. 9. (not submitted for publication).

During FY 1975 a total of 19 research projects were funded. This is in addition to

projects funded in previous years, but still in progress due to multiple year funding

periods. A list of all projects currently in progress would be rather voluminous, therefore

a complete listing of such projects in progress, describing FY 1975 research grants, may be

obtained from NSF. Total dollar support for Wind Engineering research during the last five

years is shown in Fig. 10.

II-2
Fig. 1

WIND ENGINEERING

1) Structure of Wind

a) Boundary- layer-vertical (50 to 100 meter region) and horizontal profiles, turbulence

intensities

b) Extremes and durations

c) Effects of Moisture content - rain, snow, hail

2) Wind-Wave Effects

a) Wave generation

b) Open sea structures - floating platforms, ships

c) Off-shore structures

d) Coastal structures

e) Harbors, estuaries, inlets

3) Effects on Urban Areas

a) Characteristics of Wind build-up areas

b) Influence of tall structures

c) Transport and entrainment of pollutants

d) V/STOL airports and heliports

e) Micro-thermal phenomena

4) Wind Loading on Strucutres

a) Drag, vortex shedding and separation

b) Dynamic response and stochastic analysis

c) Numerical algorithms and computer software

d) Loads on cladding and glass

e) Static vs. statistical approach

f) Permeability effects

5) Severe Storms

a) Thunderstorms

b) Hurricanes - wind velocities, wind-water interactions

c) Tornadoes - wind and pressure distributions

d) Local phenomena

e) Prediction capabilities

6) Design for Hurricanes and Tornadoes

a) Critical structures - nuclear reactor, etc.

b) Standard structures - controlled-sequence failure to protect life

c) Missile damage

d) Economic considerations

7) Full-scale Testing

a) Instrumentation development

b) Urban and rural profiles for engineering applications

c) Structure testing - fixed and portable structures, bridges, cooling towers, tall

stacks, tall buildings

8) Model Testing

a) Special wind tunnel facilities

b) Model scaling and dynamic similitude

c) Structure testing - fixed and portable structures, bridges, cooling towers and tall

stacks

d) Effects of clustering and wake interaction

II-3
9) Environmental Factors

a) Effects of structure on surroundings - entrappment of pollutants, stability of

vehicles and pedestrians

b) Effects on environmental control systems

10) Psycho-Physical Factors

a) Perception levels for motion in various structures

b) Motion tolerance

c) Psychological response to drift

11) Legal Factors

a) Pollution problems

b) Buffeting of downstream structures, etc.

c) Missile damage caused by other structures

d) Structural damage

e) Occupancy discomfort

f) Insurance rates

12) Special Problems

a) Moisutre penetration of buildings

b) Wind zoning for cities

c) Building code requirements

13) Wind Considerations in Urban Planning

14) Building Codes and Regulations

15) Boci-Economic Effects

16) International Cooperation

II-4
Fig. 2

Wind

Engineering

Research

Council

An organization for coordinating

research activity and disseminating information

on Wind Engineering problems

II-5
Fig. 3

PURPOSE

The Wind Engineering Research Council was

formed to provide a mechanism for the free exchange of information on research plans, priorities,

and programs; and to assist in the coordination of

research efforts in wind engineering.

The primary objectives of the Wind Engineering

Research Council are:

• To stimulate the initiation of, and provide the

coordination, of research efforts in wind engineering.

• Implement the dissemination and exchange of

information between research workers in the various disciplines of wind engineering.

• Delineate problems that most urgently need

research.

• Establish a National Wind Engineering Information Center consisting of a library of publications and research reports.

• Provide advice to government agencies and

other interested parties on wind research efforts

and problems.

Fig. k

ACTIVITIES

The Council organizes periodic national meetings of research investigators in Wind Engineering.

These meetings consist of reports on current re- search and discussions in or by working sub-groups

on directions for future research. The proceedings

of these national meetings are published and sent

to those individuals and organizations on the WERC

mailing list. In addition, a quarterly newsletter reporting on activities and forthcoming events in Wind Engineering is mailed to all interested parties.

The activities of WERC are supported by a grant from the National Science Foundation.

MEMBERSHIP

Because of the nature of the organization there

is no formal membership classification for participants in WERC activities. The "membership" of the Council consists of all persons and organizations, on the WERC mailing list, having an active

interest in one of the areas of Wind Engineering.

Interested parties may be included on the WERC

mailing list by submitting their name, organization,

and address to:

Dr. Richard A. Parmelee, Executive Secretary

Wind Engineering Research Council

Department of Civil Engineering Northwestern University

Evanston, Illinois 60201

Phone: (312)492-3172

II-6
Fig. 5

WIND ENGINEERING

Recognizing that the effects of wind on the worl<s of man result in extensive damage and unacceptable loss of life, the Wind Engineering Research Council was formed to encourage and coordinate research on winds and wind effects, and to disseminate information. WERC is especially concerned

with the following aspects of the wind engineering problem.

BOUNDARY-LAYER WINDS

Distribution of mean air speeds

Distribution of mean temperature Turbulence characteristics Orographic effects Urban effects Joint probability of air velocity,

temperature and humidity

SEVERE STORMS

Thunderstorms

Hurricanes

Tornadoes

Extratropical cyclones

Downslope winds

Post-disaster inspections Extreme wind statistics

WIND LOADING ON STRUCTURES

Mean and dynamic responses

Vortex-shedding excitation and

galloping

Loading on cladding and glass

Effects of architectural details Stochastic analysis

Impact of wind-driven missiles

Building code requirements

PSYCHO-PHYSICAL FACTORS

Preception thresholds of motion

Tolerance for vibratory motion

Wind criteria for pedestrian safety

and comfort

EFFECTS ON URBAN AREAS

Transport, entrainment and

entrapment of pollutant

Stability of vehicles and pedestrians

Location of V/STOL airports and heliports Location of industrial and power plants

Wind zoning for cities Micro-thermal characteristics

FULL-SCALE MEASUREMENTS

Instrumentation development

Meteorological variables

Pollutant concentrations

Wind pressures on building surfaces

Structural responses

MODEL TESTS

Wind-tunnel facilities Model-prototype similitude Blockage effects Wind characteristics for complex geometry

Structural responses

Building surface pressures

Diffusion of stacl< and automobile exhausts

SPECIAL PROBLEMS

Agricultural aerodynamics

Air-sea interactions Wind-induced noise

Wind power generation

II-7
Fig. 6

WERC ADVISORY COUNCIL

The affairs of the Council are handled by an Advisory Committee which consists of the following members:

EXECUTIVE BOARD

Chairman

Jack E. Cermak

Colorado State University

1 st Vice-Chairman

Leslie E. Robertson

Ski Ming, Helle, Christiansen, Robertson

2nd Vice-Chairman

Anatol Roshko

California Institute of Technology

Executive Secretary Richard A. Pannelee

Northwestern University

ADVISORY COMIVIITTEE

Arthur N. L. Chiu

University of Hawaii

Alan G. Davenport

University of Western Ontario

Joseph H. Golden

National Severe Storms Laboratory

Robert J. Hansen

Massachusetts Institute of Technology

George W. Housner

California Institute of Technology

Kishor C. Mehta Texas Tech University

Nathan M. Newmark

University of Illinois

Hans A. Panofsky

Pennsylvania State University

George W. Reynolds

Woodward-ENVICON Inc.

Robert H. Scan Ian Princeton University

Anshel J. Schiff Purdue University

H. C. S. Thom

Consulting Meteorologist

II-8
Fig. 7

Wind Engineering

Research Digest

Volume I - 1974

Sponsored through Grant GK-38047

National Science Foundation

Conducted in Cooperation with the

Wind Engineering Research Council

University of Hawaii

Honolulu, Hawaii

II-9
Fig. 8

TABLE OF CONTENTS

PREr ACh

1 .

• Structure of Wind

Z . Wind—Wave Effects

oJ . XjXIcCuo on UJLlJcLLl rtj-GciS

AH . Wind Loading on Structures

D . Severe Storms

D . Design for Hurricanes and Tornadoes

7

Qo

.

Model Testing

q ILIi V J.I. UlUllCil Let J. r UWX o xu1 n • 11 - Legal Factors

12. Special Problems

13. Wind Considerations in Urban Planning

14. Building Codes and Regulations

15. Socio-Economic Effects

16. International Cooperation

17. Wind Energy Conversion System

AUTHOR INDEX

INSTITUTION/AGENCY INDEX

TABLE 1 - Projects Reported and Sponsors

TABLE 2 - Sponsors of U.S. Projects

ADDRESSES OF PRINCIPAL INVESTIGATORS

Fig. 10

NSF WIND ENGINEERING EXPENDITURES

FY 1971 $ 136,300

FY 1972 630,900

FY 1973 1,013,200

FY 1974 772,800

FY 1975 $1,222,800

11-10
STUDY OF THE WIND PRESSURE AND THE RESPONSE OF ROOF CORNERS

by

Tatsuo Murota

Senior Research Officer

3rd Research Division

Building Research Institute

Ministry of Construction

and

Mitsuo Nakahara

Senior Research Officer

3rd Research Division

Building Research Institute

Ministry of Construction

ABSTRACT

In 1972, a number of prefabricated dwellings in the Nagoya district, Japan, suffered

damage from high winds caused by Typhoon No. 7220- Much of the damage to the dwellings

caused by this typhoon was the removal of the flat roofs of the buildings • It is known

that the corners of flat roofs, when subjected to high winds have high suctions with periodic fluctuating force components. The cause of this type of damage has created a need

for studying the dynamic effect of wind pressure on roof corners. Such a study has been

conducted since 1973 by the Building Research Institute. The results of the 1974 field

observations, by the BRI, will be given herein.

Key Words: Buildings; Dynamic Effects; High Winds; Roofs; Typhoon; Wind Pressure.

11-11
Full-Scale Measurement of Wind Pressure

A two-story dwelling, shown in Fig. 1, was used to measure wind pressures on roof

corners. The locations of the pressure measurements were arranged on the roof of the 2nd

story as shown in Fig. 2.

Typical power spectra of the wind pressure fluctuations, at several pressure measurement locations , are shown in Figs . 3 and 4

.

Fig. 3 shows the power spectra at the north west corner of the roof during a 11 m/s

wind from NE-E direction. At the NE-1 location, the energy power distribution occurs in all

frequency ranges and, therefore, the pressure fluctuation is considered to be random. However;

at locations NE-2 and NE-3, two predominant frequency components are observed at values of

0.4 and 0.72 H . At location NE-2, the energy peaks are observed at 1.9, 2.2 and 2.8 H , z z and at location NE-4 only a line power is observed at 2.65 H^.

The reduced frequencies are given by;

where n = frequency, D = characteristic length, and V = wind velocity.

Using the predominant frequencies of 0.4 and 0.72 at locations NE-2 and 3, the

reduced frequencies are 0.17 and 0.13, respectively, where the characteristic lengths D are

taken as 2h (h = height of the wind story roof from the 1st roof) and 21 (1 = projected

length of the eave), respectively. These values of reduced frequencies are approximately

the same as those obtained on square cylinders during wind tunnel tests , and it is

considered that these predominant frequency components are caused by the vortex formation

near the roof corner.

The physical meaning of the predominant frequency of 2.65 at location NE-4 has not

been determined and will require future studies

.

At locating NE-5 and 6, the energy peaks, as shown in Fig. 4, are also observed to

have the same frequencies as those at locations NE-2 and 3 as shown in Fig. 3. Consider,

for example, point NE-1 shown in Fig. 4, which has a peak frequency of 1.6 H^. Suppose

that D is 2 times the projected length of the curve, the reduced frequency becomes 0.18.

The power spectrum for point NE-7 has the same feature as that for point NE-4 as shown in

Fig. 3.

Model Test on the Response of Roof Corners

A small model of a roof corner was set on a building 60 m high above the ground and

the vertical displacements of the roof corner was observed for 0° to 30° roof slopes, as shown in Fig. 5. The model was set at a position where the wind direction was constant and

the vertical wind component was very small.

The results from these observations are as follows;

a) The mean displacement changed with roof slope: that is the direction of the mean

response was downward for 0° to 10° and upward for 15° to 30° as shown in Fig. 6.

11-12
b) The root mean square values of the displacement O^, was independent of the roof

slope. Assuming that the maximum response Y and the mean response Y is given

max mean

by;

Y = Y (1 + „ , max mean Y ' mean

the constant k equals 3 to 4, in the range of roof slope 0° to 30° as shown in

Fig. 7.

c) Fig. 8a to e shows the power spectra of the displacement for the roof slope 0°

to 10°. In these figures the energy peaks are observed at the fundamental frequency of this model roof. However, when the roof slope is in the range of 15°

to 30°, these peaks are not observed, as shown in Fig. 8f to h.

Conclusion

The following phenomena has been observed from the measurement of wind pressure and the

response of the roof corners.

1) The predominant frequency components were observed during the wind pressure fluctuation at the roof corners.

2) The values of the reduced frequency of these predominant components were nearly

equal to those recorded on square cylinder;^ during wind tunnel tests and these

components are considered to be caused by the vortex formation

.

3) The response of roof corners in the wind, from the diagonal direction, changes

its feature discontinuously for a slope angle of 10° to 15°.

Acknowledgment

This study was conducted at the Building Center of Japan in 1974. The authors wish

to express their thanks to the chaiirman, Dr. Isamu Kamli and the other members of the

Task Committee on this study organized in BCJ.

1

11-113
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11-14
11-15
P.H.

A/iemometer

Instruments

Model

Fig. 5 Plan and elevation of the roof-corner model

Arrangemert of the irodel on a building.

11-16
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11-18
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11-19
ON THE WIND RESPONSE OF THE KANMON BRIDGE

by

T . Okubo

Head, Structure and Bridge Division

Public Works Research Institute

Ministry of Construciton Japan

N. Narita

Chief, Structure Section

Public Works Research Institute

and

K . Yokoyama

Research Engineer, Structure Section

Public Works Research Institute

ABSTRACT

The resistance against wind of a long-span suspension bridges is one of the important

design problems because this type of bridge is flexible and sensitive to wind action. The

Kanmon Bridge, opened to the public in November 1973, is a typical long-span suspension

bridge in Japan. The design of this bridge against wind initiated during the early design

phases and its stability against wind was confirmed through wind tunnel experiments on section models. However, confirmation between theory and model tests can only be confirmed

by considering field studies and the resulting data. Such studies can reconfirm the

bridge safety when the bridge behavior is examined under strong wind conditions. Considering these thoughts, a long-term observation plan system was established on the Kanmon

Bridge, with actual observations initiated in April 1974.

In this paper, the research plan for the long-term observation system is reviewed.

Then, the characteristics of the approaching wind and the response of the suspended structure are presented. A comparison is then made between observed and theoretical wind gust

responses. The results show that the observed response is less than the theoretical response and thus shows the need for accumulation of data which can improve analytical

techniques

.

Key Words: Bridges; Displacements; field Tests; Winds; Wind Measurements; Wind Speed.

11-20
Introduction

Wind effects on a structure are generally classified into two categories, static and

dynamic effects. Generally in suspension bridge design only the mean drag component is considered as design wind load. The design wind load is generally of minor importance for

short span bridges, but is important in long span bridges, because the cross-sectional area

of the principal members in long-span bridges are governed by wind load.

In addition, there is a possibility that suspension bridges or cable-stayed girder

bridges will demonstrate aerodynamic instability such as flutter or vortex-excited oscillations as a result of the dynamic effect of the wind. Therefore, design information and

girders are required to account for this effect in the design procedure.

Then, during the preliminary design of the Kanmon Bridge, a comprehensive survey was

conducted. The basic wind speed (expected wind speed averaged over ten minutes at a height

of ten meters above sea level) for determining the design wind speed was selected based on

both the expected return wind speed at the construction site and the wind distribution

estimated from the wind tunnel experiment on a topographical model. The cross sectional

configuration of the suspended structure was selected from section model tests under laminar

flow, which showed sufficient stability against the wind.

In addition to the Kanmon Bridge, several other long span bridges such as the Wakato

Bridge and the Onomichi Bridge have been constructed in Japan. The construction of these

bridges were made possible by the results obtained during the Honshu- Shikoku Bridge Project

study.

The first wind resistant design criteria, for the Proposed Honshu-Shikoku Bridge was developed in 1967. The revised slope of this design criteria was initiated in 1972 in

conjunction with a field study. An analytical method based on statistical concepts for

buffeting the problem of suspension bridges was subsequently adopted.

The design specifications for the Proposed Honshu-Shikoku Bridge (1975) provides for

a design wind speed (V^^) and a design wind load (P^^) which are computed by the following

equations

:

for horizontal line-like structure: V^^ = Vi*V2*Vio

for vertical line— like structure: V^.^ — Vi*V3*Vio

for horizontal line-like structure: P^ = l/2pv«C„*An'Vit

D d D

for vertical line-like stracture: P = l/2py,'C 'An'Vs

D d D

where

,

basic wind speed (m/sec)

Vi : wind speed modification factor

V2 : wind speed modification factor

V3 : wind speed modification factor

P : 2 -4

air density (Kg 'sec 'm )

for the elevation of the structure

for horizontal length of the structure

for vertical length of the structure

11-21

,
C : drag coefficient

D

2

An: exposed area for unit length of the structure (m /m)

Vi+ : wind load modification factor for horizontal length of the structure

V5: wind load modification factor for vertical length of the structure.

The fundamental concept used in this design criteria is to determine the wind speed

modification factors and the wind load modification factors such that the magnitude of the

lateral bending moment of the stiffening trusses caused by the natural wind is equal to

that induced by a uniform wind. Thus analytical method is based primarily on the statistical concepts proposed by A.G. Davenport in 1952. However only a few suspension bridges

have shown lateral vibrations during stoinny winds, thus this infrequency has been introduced in the analytical process. In order to further substantiate these assiamptions field

observations on full scale structures and wind tunnel tests on section models, have been

conducted. A long-term observation system was then installed on the Kanmon Bridge, in

order to collect useful data and thus confirm the design adequecy of the bridge and enhance

the design philosophy of the proposed Honshu-Shikoku Bridges.

The Kanmon Bridge and its Aerodynamic Characteristics

The Kanmon Bridge, which was constructed over the Kanmon Straits and forms the

Kainmon Highway- connects Shimonoseki City in Honshu and the Moji District of Kitakyushu City.

This is the largest suspension bridge in Japan, and has a center span length of 712 meters

with equal side spans of 178 meters in length. The suspended structure carries six traffic

lanes and consists of stiffened trusses, nine meters deep spaced at 29 meters. The clear

height is 61 meters above sea level at the center span. The cross-sectional shape of the

stiffening girder was selected through wind txannel tests. The elevation of the upper chord

of the stiffening truss is nearly equal to that of the road surface, with three lanes at

the edge and central part of the road deck, constructed of open grid. These constitute a positive characteristic of the bridge relative to wind stability.

The location, general view and fundamental dimension of the bridge are shown in Figs.

1 and 2 and Table 1.

The wind speed survey and the wind tunnel model tests, were conducted in relation to

the wind resistant design. In the former, the vertical profile of the mean wind speed, and

the vertical inclination of wind were investigated. Also studied in detail was the relationship between the wind direction and the wind convergence as the topographical conditions of

the construction site was modified. As a consequence of this study, the wind distribution

was determined and the basic wind speed was fixed at 35.5 m/sec, which then gave a design

wind speed as 53.5 m/sec for the stiffening girder. In the latter, the mean wind force and

wind stability of the suspended structure were then examined. Two types of stiffening

girder were proposed for the suspended structure namely, the box-stiffened girder and the

truss-stiffened girder. After prudent comparison of the wind stability for both systems,

the truss-stiffened suspended structure was adopted. The wind stability, of the final design structure, is given in Fig. 3 and indicates the relationship between the angle of

11-22
attack and the critical wind speed for the torsional flutter. The occurrence of the restricted oscillation was anticipated by the wind tunnel test on a main tower model under

construction, and the oscillation was suppressed by a special energy absorption device.

The design wind load for a suspended structure and the main tower was determined in

accordance with wind tunnel test results.

After completion of the actual bridge structure a field vibration test was conducted

by use of twin exciters and heavy trucks . The natural frequencies and the damping characteristics obtained from the test are given in Table 2. The observed frequencies are about

four percent less than those calculated, which suggests that the critical wind speed should

be decreased by the same percent. Though the observed damping capacity was smaller than

that expected, it does not appear necessary to modify the critical wind speed by the

magnitude, because the observed amplitude of the oscillation was small compared to the

expected value.

Long-term Observation System

An instrumentation and data-recording system, designed to monitor and record wind

speed and direction as well as earthquake tremor and to indicate their effect on the

structure, has been provided by the Japan Highway Corporation. The location of these instruments are shown in Fig. 4 and Table 3. The characteristic features of the system

are (i) a large member of instruments are provided, (ii) the system is immediately available to record and monitor, and (iii) a digital signal is used in the transmission, in

order to increase the accuracy.

(1) Transducers

(a) Propeller type anemometer: Four anemometers were positioned on the top of

poles located on the bridge deck, and used for observing the horizontal distribution of the wind. One anemometer was installed at the top of the main tower

and used as reference.

(b) Sonic anemometer: One set of three-component sonic anemometers was located

at the mid-point of the main span and used for observing the wind inclination

angle.

(c) Servo- type accelerometer : Servo-type accelerometers of high fidelity were positioned at the mid and quarter points of the center span. One element mentioned

the horizontal component and the other two elements the vertical component, at each

point

.

(d) Displacementmeter : Using an optical device, both horizontal and vertical

motions of the stiffening structure were detected.

(e) Others: Transducers used for earthquake detection were placed on the

structure, but are not described herein.

(2) Data transmission system

The output data are digitized by Analog-to-Digital converters near the transducers

and are transmitted 1.5 kilo-meters to the data processing system located in the

administration office.

11-23
(3) Data processing system

A data processing system equipped a mini-computer ordinarily performs the monitoring and records the data. When the wind speed exceeds a preset level, the acquisition

system automatically starts and the data is stored on magnetic tapes.

Observed Data and Discussion

Wind Characteristics

The observation system was set into operation in April 1974. As strong typhoons have

not been in the ara thus far, only a few recordings at low wind conditions were obtained.

Wind characteristics have been examined in order to compare the real wind distribution

with that expected. Fig. 5 indicated a trace of the wind speed and the direction, at four

stations. From this information statistical data such as mean wind speed and standard

deviation are calculated as shown in Fig. 6 which gives the variation of the wind with time.

A set of power spectral density is given in Fig. 7, from which the characteristic features

of the approaching wind is

:

(1) The turbulence intensity, gust factor and power spectral density of the wind

agree well with those obtained previously in the past, with a noticeable effect of

wind convergence.

(2) The horizontal distribution of the wind speed was observed on the deck and is

almost uniform, with a small deviation of 1 m/sec. The mean wind direction is nearly

equal for each station. However, instantaneous wind speeds and directions fluctuate

so violently, that cross-correlation of the wind speed rapidly decreases with an

increase in the distance between stations. According to the cross-correlation coefficient of wind speed as shown in Fig. 8, the wind speed has a strong correlation with

the anemometers set at 20 meters apart, but has a weak correlation for those meters

which are 180 meters apart.

(3) According to the topographical model tests, the vertical inclination of the wind

at the straits varies with the principal wind direction. For instance, the S-W wind,

which comes from Mt. Hinoyama (EL. 268 m) , has an angle of inclination of 3 degrees at

the Shimonoseki side, with the N-E wind showing a negative angle on that side.

The recordings of the sonic anemometers at the east side of the bridge deck showed a positive angle for both the E-SSW wind and the E-NNE wind.

Response of the Stiffening Girder

In order to better understand the behavior of the suspended structure when subjected

to wind, a motion picture was made by a computer output microfilming system. This motion

picture showed that the bridge vibrated randomly in vertical bending, torsional, and

lateral bending modes with the vertical bending mode being dominant. Fig. 9(1) through

Fig. 9(5) shows the power spectral density of the acceleration for each mode, and Fig. 9(6)

and Fig. 9(7) displacement mode. In Fig. 9(1) and Fig. 9(2) the natural frequencies are

indicated, which were obtained by forced vibration test. These frequencies coincide with

the peak frequencies of the power spectral density curve. However, the peak frequencies

of the lateral bending vibrations are smaller than the calculated frequencies using the

Ellis' method. It was determined, from the acceleration record, that the response vibra11-24
tion of the suspended structure is random in character and the lowest mode of either symmetry or antisymmetry is dominant. According to the power spectral density of the displacement the contribution of the higher mode of vibration is smaller than the lower mode of

vibration. It may be said that the vibration with the lowest mode is important not only

for the flutter phenomena but for the gust response problem.

Wind Speed and Structural Response

In Fig. 10 and Fig. 11 the original trace of the wind speed and the sturctural response

is given. The structure starts to vibrate at a wind threshold level during the combined

vertical and lateral bending mode and is random rather than sinusoidal. The magnitude of

the vibration seems to depend on the wind speed level rather than on the wind turbulence

level, and with an increase in the wind speed the lowest mode of the vibration becomes

dominant

.

Fig. 11(1) shows experimental and analytical mean and the standard deviation of the

lateral displacement at midpoint of the center span in relation to the transverse component

of the mean wind. The analytical method that was used is the same as reported at the last

Joint Meeting. According to Fig. 11, the observed values coincide qualitatively with the

theoretical value but the observed mangitudes are smaller in comparison to the theoretical

values. This indicates the necessity to reexamine the wind characteristics (spatial distribution, power spectral density of wind) , structural characteristics (natural frequency,

damping capacity) and the aerodynamic response characteristics (aerodynamic admittance,

aerodynamic damping)

.

Concluding Remarks

The main part of this work has been concerned with the examination of data collected

by a long-term observation system, in which only a few samples of data have been obtained.

The primary analysis of this minimum data indicates the following;

(1) The observed wind characteristics (power spectral density of wind speed, turbulent

intensity, and gust factor) are similar to those observed previously. The horizontal

distribution of the wind speed and the direction is nearly uniform, however, the wind

convergence at the straits has not been confirmed. The wind inclination data shows

that reverse characteristics in comparison with wind tunnel experiments, thus indicating a detailed survey is necessary.

(2) During the field studies no information of the flutter problem was obtained.

However, some response characteristics of the suspended structure, against natural

wind, were obtained. It was found that the structure vibrates randomly and that the

higher mode components as well as lower mode ones are included in the power spectral

density of acceleration but the lowest mode is dominant in displacement.

(3) The observed vibrational displacement shows the same tendency when compared to

the calculated values based on Davenport's method, but does not agree quantitatively.

11-25
Acknowledgment

The authors are indebted to Mr- M. Hurumichi of Japan Highway Corporation who assisted

during the field work. The authors also wish to thank Mr. M. Katsuragi and Mr. K. Nomura

of the Structure Section P. W.R.I, for use of the computer program.

11-26
References

1. Meteorological Agency, Ministry of Construction: Investigation into the Second Kanmon

Road, March 1967

2. Soma, S.: Wind Tunnel Test of Air-stream at Kanmon Straits, November, 1967

3. P.W.R.I.: An Investigation for Design and Construction of the Kanmon Bridge, March

1973, P. W.R.I. Tech. Note

4. P.W.R.I.: Report of Vibration Test on the Kanmon Bridge, March, 1974, P. W.R.I. Tech.

Note.

5. Narita, N. and Yokoyama, K. : On the Gust Response of Long-span Suspension Bridge

(Progress Report), May, 1974, Sixth Joint Meeting, UJNR, Washington, D.C.

II-i27
B

•HQ

Side

span

178.0

167.75

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25.946

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2.916

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Center

span

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11-28
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11-29
Table-3 Instruments for long term observation system at Kanmon Bridge

Observat ion

item

Measuring

point

Maximum

measuring

range

Instruments Number

of

instruments

Number of

elements Rate of

sampling

(f req . / sec) per

M.P. Total

Acceleration

Ground ±500 gal Submerged

accelerometer

3 3 9 100

Abutment ±500

Electromagnetic

accelerometer

4 1 4

50 Pier ±500 8 1 8

Tower ±500 4 1 4

Stif f enin

frame

' ±200 Servoaccelerometer 6 1 6 25

Displacement

Stiffening

frame

Horizontal

±2 m

±2 m

XY-Analyser 1 3 3 25

Longitudinal

±25 cm

Differential

transducer 1 1 1 25

Wind

speed

Stiffening

frame

Horizontal

60 m/sec

Vertical

±10 m/sec

Supersonic anemometer

1 3 3 25

Wind

direction

Horizontal

60 m/sec

540°

Propellar type anemomer

3 2 6 25

Wind

speed

Top of

Shlmonoseki tower

Horizontal

60 m/sec

1 2 2 25 Wind

direction

540°

11-30
Fig.-l Map of the Kanmon Straits

11-31
11-32
-4 -3

Angle of attock I Degree I

Fig. -3 The aerodynamic characteristics

of the Kanmon Bridge

11-33
Q <44rt

1

1 f

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11-34
f

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11-35
OMP-A AMP-C

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9

A

A

o

1 0 11 12 13 14 (Time)

&

&

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A

i

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13

ti.

14 (Time

)

o

A

%

8

o

—I

11 —r1 0 12 13 14 (Time)

o A

6

A

A

o

10 11 —I12 13 14 (Time)

Fig. -6 The variation of wind characteristics

with time (18 Nov. 1974)

11-36
c cs/n/2»»(s/w) ) xiisN3a nvai03ds asMOd

( CS/t)/Z»»(S/W) ) AilSN3a lvai33dS a3W0d

11-37
( C9/l)/2»»C9/W) ) AilSN3a "lvai03dS a3M0d

CPOSS-CORRELAT i OM COEFF I C i EI-iT

1 .0 T

Fig. -8 Cross-correlation coefficient

11-38
I

11-39
ii-4o
11-41
POWER SPECTRUM OF RESPONSE

1974. 1 1.18

8.50.00

1 0 -J

1 1 r—[ I r I I

I

1

r—i—[ r r i i |

1 1 1—| i i i i

| -2 -1 0 ' 10 10 10 10 FREQUENCY (HZ)

Fig. -9 (7) Power spectrum of Response

(M.P.-A. Horizontal)

11-42
11-43
18 Nov , 21 Nov 1974

Hwiiontol disploctmeni Mean

{cm) (cti

50 lOH

Dm,on (Theoretical) ^ ^^^^ ^^^^^ horizontol

displacement

•^0 ( Theoretical ) ^. ^ ^ ^ • . X Stondord deviatioft

of horizontol displacement

20 30 Meon wind speed perpendicular to bridge oxis (MP -A)

Vertical displacement

Fig. 11(1) Wind speed v.s. Horizontal displacement

XX

X : Standard deviation

of vertical displacement

Cn/s)

20 3^ Mean wind speed perpendicular to bridge axis (MP. -A)

Fig. 11(2) Wind speed v.s. Vertical displacement

11-44
U 0)

4-1 -H

O) CO EO -H

V4 ^

CU 01 T—I cn

<u o a C o o

« 6H

U ^ •H Cn

4-1 0)c

CO

to u

f u

o o

U 4-1

4-J U <D 0) ^

CO

I

Lerometer

-component

!

Servo-acce]

(M.P.-A,

V-

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11-45
11-46
LULING, LOUISIANA CABLE-STAYED BRIDGE

WIND TUNNEL SECTION MODEL TESTS

by

Richard H. Gade

Research General Engineer

Office of Research

and

Walter Podolny, Jr.

Structural Engineer

Office of Engineering

and

Harold R. Bosch

Structural Research Engineer

Office of Research

Federal Highway Administration

Washington, D. C.

ABSTRACT

Results of wind tunnel section model tests of five orthotropic superstructure configurations for the Luling, Louisiana, Cable-stayed bridge are presented.

The 1,235-foot (376.4 meters) long main span crossing of the Mississippe River is

designed for 150 miles per hour (67 meters per second) hurricane wind velocity.

The section models, 1/60 scale, are tested at six wind angles: -4°, -2°, 0°, 2°, 4°,

and 6°. Flutter coefficients are plotted for all tests.

The tests show freedom from flutter at the design wind speed. Vortex excitation,

vertical and torsional, is exhibited.

Testing is performed in smooth flow conditions at the Fairbank Highway Research

Station of the Federal Highway Administration, utilizing the George S. Vincent Wind Tunnel.

Key Words: Bridge, cabled stayed, wind tunnel, models, displacements, flutter, wind

angles

.

11-47
Introduction

Recent bridge structures have increased dimensions and flexibility and decreased dead

weight and damping characteristics. Reduction of dead weight produces a magnification of

wind effects relative to the inertia of the structure. Increased flexibility decreases the

natural frequency of vibration. Modern fabrication techniques have decreased the structure's

ability to absorb energy by sliding friction between component parts and thus less energy

is required to initiate and maintain vibration. As a result, recent structures are becoming more sensitive to not only static wind effects but dynamic ones as well. Some ex- isting and relatively recent structures have been so affected by wind oscillations that they

have required reinforcement or modification by fairing.

The assessment of dynamic deflections and oscillations of long-span bridges due to

wind action usually resorts to a wind tunnel model test. This is particularly so for the

traditional truss stiffened suspension bridge. The cable-stayed box-girder bridge which is

currently receiving popularity by designers in the United States is subject, fundamentally,

to the same wind excitation mechanism as the traditional suspension bridge. The inherent

general increase in stiffness of the cable-stayed box-girder does place it in a different

realm of response.

This paper will present data gathered from model studies conducted by the Federal

Highway Administration (FHWA) on the Luling, Louisiana cable-stayed bridge.

Wind Force and Angle of Attack

Wind force on an object is normally not in line with the direction of the wind. In

conventional aerodynamic analysis, i.e., airfoil design, wind force is divided into two

components: drag and lift, parallel and perpendicular to the wind direction. This same convention may be applied to a bridge deck wherein the resultant wind is oriented to the

structure by the angle of attack a, positive when striking the section from the underside.

It is convenient in considering wind effects on bridge structures to consider lift being

perpendicular to the normal cross-section position of the bridge deck.

In the evaluation of wind forces on a structure there is a concern for the determination of the possible direction of critical wind velocity. In plan it is generally assumed

that the critical wind direction is perpendicular to the longitudinal axis of the bridge.

An obvious question arises as to what maximum value of the angle of attack, a, should be

considered on the deck cross-section. The angle of attack is a function of wind velocity

and site conditions. Preliminary data for the relationship between maximum observed angle

of attack and wind velocity obtained by the FHWA at the Newport, Rhode Island Suspension

Bridge over Naragansett Bay (Fig. 1) provided measurements of the mean vertical wind angle

for a range of wind speed recorded for Hurricane "Doria". Form these curves it can be seen

that the angle of attack decreases with increasing wind speed, thus, at lower wind speeds

the structure must be stable for larger values of angle of attack. The curves shown in

Fig. 1 may serve as a guide but may not necessarily be applicable to other sites and

possibly may impose unnecessary constraints to the analysis.

11-48
Luling Bridge

Wind tunnel tests are being conducted on 1:60 scale section models for the proposed

bridge deck superstructure cross-section shapes of the Luling, Houisiana, cable-stayed

bridge over the Mississippi River. These tests, still in progress, are being conducted at

the FHWA's Fairbank Highway Research Station, McLean, Virginia, using the low speed, smooth

flow, 6 ft. by 6 ft. open jet, George S. Vincent Memorial Wind Tunnel facility.

The experimental procedure. Fig. 2, suspends the essentially rigid model by four equal

springs at each corner of the end supporting brackets. End plates are employed so that

the air flow near the end of the model is two-dimensional. The models, five feet in span,

are scaled by the Froude Number Criteria for mass property and vertical natural frequency.

Due to practical limitations of simulating the relatively high prototype torsional characteristics, values of polar mass moment of inertia and torsional natural frequency do not

achieve Froude Scaling.

The behavior of damped spring-mass systems is dependent on the value of the damping

coefficient. A specific value of structural damping for a proposed bridge structure

(which for practical purposed is impossible to obtain) is difficult to assign, and consequently, so is the value for a model system. The section model testing is, therefore,

treated as a method of extracting aerodynamic information which can be used analytically

with any selected value (or values) of prototype structural damping. Thereby, the need for

adjusting the model system to a specific value of structural damping is avoided. The tests

are conducted with the values of mechanical damping exhibited by the freely oscillating

model.

Seven superstructure configurations are considered, all having a center line median

barrier. A typical section model of the double trapezoidal box-girder design with an orthotropic deck is indicated in Fig. 3. The same section model modified by a slanted, non

structural, fascia plate is shown in Fig. 4. Initial wind tunnel tests suggested that this

"faired" cross-section would exhibit a marked improvement in aerodynamic response. A section model of a multi-cell steel box-girder of a "streamlined" cross-section is shown in

Fig. 5.

The aerodynamic information obtained from the section model tests is in the form of

flutter derivatives, following the procedure and linearized treatment of Scanlan and Tomko.

The mathematical model for lift and torsional moment are indicated in Eq. (1)

.

+ KH

2

*

(1)

11-49
where a angular displacement

B deck width

* Hn . r 1* A.

X \' damping ratios respectively in vertical and torsional motions of the

free bridge

\' 03a

= corresponding natural frequencies

m, IP mass and mass moment of inertia of the deck per unit span, referred

to the elastic axis

h

_ vertical displacement

p air density

V wind velocity

VIN,

0) circular frequency of oscillatory motion

The Luling Bridge models exhibited uncoupled response motion, therefore, only the

H^, A* and A* stability derivatives are applicable. is the function for aerodynamic

damping of vertical oscillation, A* is indicative of torsional aerodynamic damping and

* A^ is the measure of change is torsional frequency between aerodynamic damping and mechanical damping.

Six wind angles varying from -4° to +6° and increments of equivalent prototype wind

speed to 160 mph were investigated. The experimental values for the respective derivatives

are plotted as functions of reduced velocity (v/NB) which may be considered as a non- dimensional frequency.

An example of the test results are shown in Fig's. 6 through 8 for H^, A^ and A^. The

test results illustrated are for model C-2C-A at zero wind angle.

All model shapes showed freedom from self-excited, divergent, vertical flutter oscillation. Torsional flutter occu at wind speeds well above the design value of 150 mph,

or at unlikely wind angles. However, all models responded, variously, to vortex shedding

response. This is an amplitude limited response whereby the periodic shedding of vorticies

in the wake of the structure is resonant with the vertical and/or torsional modes of the

structural system. This amplitude limited oscillation can be an unacceptable characteristic

of the design when it occurs at moderate wind speeds and the resulting acceleration of the

oscillating structure is disturbing to the user. A small scale (1:150) sectional representation of the cross-section scheme C-2C was placed in an air flow containing smoke filaments for purposes of flow visualization. Flow

separation tripped by the leading parapet railing, and again by the median barrier is

illustrated in Fig. 9(a). Fig's. 9(b) and (c) capture the trailing vortex.

An estimate of the prototype response to vortex shedding is provided by data from the

3 model tests. As suggested by Scanlan , the analytical model for the homogeneous selfexcited condition of "flutter methodology" is modified by providing a special periodic

forcing function applicable at the Strouhal or vortex exciting condition. For the case of

purely vertical response motion, the test observations provide a derivative, H*, characterizing the decay from an amplitude greater than resonance to the steady-state resonant

11-50
amplitude, and a dynamic lift coefficient C , Eq. (2). With these coefficients, and a

LjO

selected value of prototype damping, a conservative estimate of prototype behavior can be

calculated, when 100% lateral coherence of vortex exciting wind flow is assumed.

m li + 2 h + (Ah = pv^) (2B) (2)

A siammary of the test results for three comparative models is shown in Fig's. 10

through 12. The oscillating response characteristics of the section shape, referred to

equivalent prototype wind speed, and for wind angles of attack are plotted. The shaded

bands show the "locking in" range of wind speed over which vortex excitation takes place.

The natural wind angle boundaries are for the Severn Bridge reference data, and for the

recorded data on the Newport, Rhode Island suspension bridge.

For the same sections, an estimation of the double amplitude response of the prototype structure to vortex induced oscillations have been calculated and plotted in Fig's.

13 through 15. The ordinate scale is critical damping ratio. The abscissa scale is feet

of double amplitude at center of span with associated value of root mean square g's of

acceleration.

Closure

As a result of innovations in structure concepts, sophisticated structural analysis,

materials, fabrication and erection procedures, recent bridge structures are becoming more

sensitive to not only static wind effects, but dynamic ones as well. As a consequence,

wind forces have taken on an increased importance and significance and can be a major problem in cable- supported bridge systems; serious consideration of these forces is required

by the designer.

While most problems of bridge stability resulting from wind forces are recognized, the

acquisition of data for new designs in the wind tunnel is mandatory to produce adequate

wind stability information to enhance their design. At the present time there is no purely analytical methodology to preclude wind tunnel testing.

Data has been presented on current or on-going studies of the aerodynamics of selected

structures. The method employed of extracting the aerodynamic derivative data from test

results places no A Priori restriction on the frequencies, damping value, or inertias em- ployed in the section model although it is suggested that highly unlikely values not be

used.

It is important to note that while the validation of stability of the completed

structure (for expected wind speeds at the site) is mandatory, it does not necessarily

imply the most critical stability condition of the structure. A more dangerous condition

may be during erection when full connection of the joints may not have been established and

thus full stiffness of the structure is not achieved, the frequencies are lower than in the

final state and the ratio of torsional frequency to flexural frequency approaches unity.

The use of welded components in towers has contributed to the susceptibility of tower vibration during erection.

11-51
Component parts of the structure are also in themselves susceptible to wind excitation.

The calbes of cable-stay bridges, the hangers of suspension bridges and arch bridges, and

the towers of suspension and cable-stay bridges have been known to present problems, usually

from vortex excited vibrations.

Consideration as to an acceptable level of motion falls into two catagories: 1) structurally damaging motion and 2) human response motion. The first relates to violent motion

that may be catastrophic and/or motion that over a period of time may lead to fatigue related failures. The second relates to motion that may not be structurally damaging but may

be objectionable from the stand point of user acceptance, i.e., vibration that is noticeable

to pedestrians or occupants of standing or moving vehicles.

A great amount of research needs to be expended to study the stability of bridges of

new designs, to develop criteria to rectify existing bridges that are or may become

troublesome from a wind stability point of view, and to develop adequate procedures such

that designers can produce stable designs at the conception of a bridge or at least to minimize the probability of unstableness.

A necessary prerequisite to any attempt at developing design criteria and/or predictive

techniques that will produce a reasonable confidence level of aerodynamic stability for

various cross-sections in use or contemplated and the accommodation of the vagaries of the

natural wind will require extensive additional field and laboratory work.

11-52
References

1. Roberts, G. , "Severn Bridge - Design and Contract Arrangements," Proc. of the Inst. Civ.

Engrs., Vol. 41, September 1958.

2. Scanlan, R. H. and Tomko, J. J., "Airfoil and Bridge Deck Flutter Derivatives," Journal

of the Engineering Mechanics Division, ASCE, Vol. 97 No. EM 6, December, 1971.

3. Scanlan, R. H. , "Vortex-Shedding Response," Private Communication, R. Scanlan to R. Gade,

1974.

4. Wardlaw, R. L. , "A Review of the Aerodynamics of Bridge Road Decks and the Role of Wind

Tunnel Investigation," U. S. Department of Transportation, Federal Highway Administration, Report No. FHWA-RD-73-76

.

5. Bosch, Harold R. and Gade, Richard H. , "Wind Tunnel Section Model Tests, Luling, Louisiana, Cable-Stayed Bridge," Interim Report, Federal Highway Administration, Washington,

D.C., April 1, 1975.

11-53
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11-54
11-55
11-56
Luling Bridge Section Model C-2C-A Fig. 4

Luling Bridge Section Model C-6-C-B

11-57
0.00 1.00 2.00 3.00 4.00 5.00 6.00

V/NB

EXPERIMENTRL RESULTS FOR HSl

LULING LOUISIfiNfi CflBLE-STflYED BRIDGE

MODEL C-2C-fl Fig. 6

11-^8
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EXPERIMENTni RESULTS FOR RS2

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Fig.

11-59
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EXPERIMENTRL RESULTS FOR nS3

LULING LOUISIANA CABLE-STATED BRIDGE

MODEL C-2C-A

Fig. 8

11-60
SMOKE FLOW VISUALIZATION MODEL

11-61
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11-62

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11-54
11-65

11-67

THE REGIONAL DISTRIBUTION OF THE EARTHQUAKE

DANGER IN JAPAN

by

S. Hattori, Y. Kitagawa and T. Santo

International Institute of Seismology

and Earthquake Engineering

Building Research Institute

Ministry of Construction

ABSTRACT

The values of m and log k in Ishimoro-Iida' s statistical Formula are derived for each

component of the maximim displacement amplitude observed at many stations of the network

maintained by the Japan Meteorological Agency (JMA)

.

A distribution map of expected maximum displacement amplitudes for the earthquake

recurrence period of 100 years is made based on the derived values of m and log k. The map

indicates quantitatively the regional distribution of earthquake danger in and around Japan.

It is seen from this map that the general level of the earthquake danger varies throughout

Japan. This variation is also found to reflect the pattern of seismic activity throughout

the area. The earthquake danger increases along the Pacific side of Hokkaido, Tohoku and

Kanto districts and decreases in the southwestern and inland areas of Japan. Small variations are also recognized, which might suggest that the eaithquake danger is affected by

local geological and subsoil effects.

Key Words: Earthquakes; Earthquake Distribution; Field Data; Frequency Maps; Ground

Displacement; Japan.

III-l
Data and The Method of Analysis

The data were obtained from the monthly reports of JMA (Japan Meteorological Agency)

from Jan. 1967 to Jan. 1972, which give the relevant informations on a single earthquake

recorded at a particular station. The data are recorded on cards for each station, with the

number of cards for each year as listed in Table 1. These data are then transferred to and

stored on magnetic tapes.

Seismicity in a certain area can be measured by a well-known coefficient b in a formula

of log N = a + bM, in which N is the frequency of the occurrence of earthquake with mangitudes of M + Am, where a and b are constants.

In addition to this equation another useful formula has been derived by M. Ishimoto and

K. lida, and is;

n(A)dA = kA'^^dA (1)

where n(A) is the number of earthquakes with maximum displacement amplitude, A + dA. The

two constants m and k in equation (1) have often been calculated by means of the least

squares method. Recently T. Utsu [1] has reported that the coefficients in equation (1) can be calculated with high accuracy by means of the maximum likelihood method as expressed in

the following formula:

S4oge

(2)

. Z logAj - S logAj

where S denotes the total number of data used and A_, is the i-th maximum displacement

amplitude with the order number i from the largest value and A^ is the least value of the

maximum displacement amplitudes in the data used.

The value of the coefficient k, or log k, can be calculated from the following

formula;

Logk =logN(A :^Aj) + log(m - 1) H- (m - l)logAj (3)

where N denotes the total number of maximiam displacement amplitudes which exceed the

selected level. A..

J It must be noticed that the values of m and log k are dependent on the value of A_.

.

For instance, if too small value is selected for A_. , the number of data which may be used

becomes smaller than the actual number owing to the limitations in the ability of observation and reading of records, which leads to situations which underestimate the value of

m. Taking this into account, the data that was selected satisfied equation (1) for calculating m and log k at each station.

III-2
Numerical Analysis

In the present study, the values of m and log k were calculated from the data using

three components N-S, E-W and U-D, and resultants at 109 stations. In Figs. 1 and 2,

the relationship of (1) cumulative frequency N versus log A and (2) N versus log A^ for the

for the range of log A_. = 0.1 to 3.0, are shown for stations, Hikone, Okayama and Tokyo.

The total number of the maximum displacement amplitudes for the N-S component given in

Fig. 1(a), is 443. Because data in the range of A^ = 0.1 to 1.0 is small, the values of m

for this range are not constant and thus were not used. The trend of the variable m relative to E-W, U-D and the resultant components, has similar trends when compared to the

N-S component. Consequently in the present paper, the average value of m and the standard

deviation were all calculated from the data using the N-S component, using tentative

values at each station in the range of log A^ = 1.0 to 2.0 at intervals of 0.1. The average

value of log k and the standard deviation were then calculated using Eq. (3) . The average

values of m are shown by small open circles indicated by arrows in figures 1 and 2 and as

seen in Figs. 1(a) and 1(b) the trends are quite similar at Hikone and at Okayama.

Examination of the Tokyo Station data shown in Fig. 2, however, is quite different. In

this case, the values of m do not converge, even for a range of log A. beyond 1.0. Quite

reasonably for this case, the values of the ciomulative frequency distribution N does not

show a linear relationship relative to the log A. Average values of m and log k, at this

station have much larger standard deviations than in other cases. In Table 2, examples of

the values of m and log k and their standard deviations are listed for all 109 stations.

In this Table Nl, N2 and N3 respectively means the total number of data, the number of

data for log A >_ 1.0 and for log A >^ 2.0. Fig. 3 shows the differences between the values

of m and log k for different components at approximately 60 stations. As can be seen from

this figure, some stations have a value of m equal to 1.2 or 1.4. These exceptional small

values, however, were due to an insufficient number of data. Examination of the calculated

values of m and log k using different components, showed that there was indifference from

station to station, as mentioned previously.

Fig. 4(a) shows the regional distribution of the parameter m calculated from the

data on N-S components, while in Fig. 4(b), the distribution of m excluding those of

values of m less than 1.5. As can be seen in Fig. 4(b), the stations at which large values

of m were observed, tend to be distributed near the most active violent seismic zones in

Japan. These regions are along the Pacific coast from Hokkaido and from Tohoku through

the northern part of Kanto down to the southern parts of Chugoku and Kinki districts.

Another large zone where m is large, is the Japan Sea along the Chubu district. The locations of the various districts and stations are shown in Fig. 5. The general tendency of

the distribution of m have been found to coincide with the regional distribution of b-values

which were published previously [3] . The distribution of log k parameter is shown in Fig.

6.

The frequency T (in years) , when the stations experience a maximum displacement

amplitude in a certain period, were calculated using the values of m and log k as follows;

III-3
Recalling that Eq. (1) shows that the variable k gives the frequency of the shock and

the displacement amplitude A of 1 micron at a certain station, where this frequency is

considered to be proportional to the period of the observation. Under such an assumption,

the value of k for the period of T years can be extrapolated by using the following

formula;

61

where k.. means the value of log k for 61 months. This means

61

logk =k6, + log(12T/61)

and for this paper, the value of k for T = 100 years will be derived. Examine now Fig. 7,

which shows the regional distribution of the frequency when the maximum displacement

amplitudes of 100 micron are expected in 100 years. Figs. 8(a), 8(b) and 8(c) show the

regional distribution of the frequency, when the maximum displacement amplitude will exceed 100, 1,000, and 10,000 microns. For instance, as shown in Fig. 8(a), station Nemuro

is expected to experience 2030 ground vibrations with a maximum displacement amplitude of more than 100 microna 100 years, which is about 20 times in one year. Similarly, the same

station will in one year, experience three and 0.4 ground vibrations with a maximum displacement amplitude of 1,000 and 10,000 microns, respectively.

Conclusions

In addition to the expectancy maps given herein, other expectancy maps, as shown in

Fig.s 9(b), (c) , and (d) have been developed by H. Kawasumi [4], I. Muramatsu [5] and K.

Kanai et al [6] . Comparing these figures with those presented herein, the following trends

can be concluded; (1) although the data was taken from different periods, and with

different accuracy, Figs. 9(a), (b) and (c) show a similar tendency in that, the earthquake danger increases along the Pacific coast of the Hokkaido, Tohoku and Kanto districts.

This reflects the general pattern of seismic activity in and around Japan. However, in Fig.

9(d), the earthquake danger in Tohoku district, appears low; (2) As shown in Fig. 9(a),

an additional area with high earthquake danger appears in the southern part of Kyushu,

(3) high values of the earthquake danger, which seem to be abnormal, appear in Aomori

(A =90cm) , Akita (A =60cm) and Nagano (A =50cra) districts as shown in Fig. 9(ai max max max

In Fig. 10, the variables N vs log A and m vs log A^ are plotted for three stations, in

order to check the reason for these abnormal values . With regard to m vs log A^ at Aomori

and Nagano stations, the values of m increase gradually in the range of log A=1.0 - 2.0

and are approximately constant for the log A greater than 2.0. As was described previously,

III-4
m was measured at all stations in the range of log A = 1.0 - 2.0. Therefore, the values

of m are underestimated at these two stations. As seen in Eq. (3), an overestimated

value of A is derived by an underestimation of m, which is why A is obtained at Aomori max

and Nagano. In the case of Akita, however, the local ground condition should be taken

into account.

III-5
References

1. Japan Meteorological Agency, 1967.1 - 1972.1, the Seismological Bull, of the Japan

Meteorological Agency.

2. Utsu, T. , A Method for Determining the Value of b in a Formula log n = a - bM

showing the Mzgnitude-Frequency Relation for Earthquakes, the Geophys. Bull, of the

hokkaido Univ., 13 (1965), 99 - 104.

3. Hattori, S., Seismicity in and near Japan - Regional Distribution of b values in log

n = a - bM, Bull USEE, 12(1974), 59-82.

4. Kawasumi, H. , Measures of Earthquake Danger and Expectancy of Maximum Intensity

throughout Japan as inferred from the Seismic Activity in Historical Times., Bull.

Earthq. Res. Inst., 21(1951), 469 - 481.

5. Muramatsu, I., Expectation of Maximum Velocity of Earthquake Motion with in 50

years through Japan, Sci. Rep. Gifu Univ., 3(1966), 470 - 481.

6. Kanai, K. and Suzuki, T. , Expectancy of the Maximum Velocity Amplitude of Earthquake

Motions at Bed Rock, Bull. Earthq., Res. Inst., 46(1968), 663-666.

III-6
Table 1 The Number of Data Cards

Year Number of Cards

1967 11,448

1968 24,192

1969 14,796

1970 11,718

1971 11,232

1972 972

Total 74,358

Stat i on c omp N 1 N2 N 3 m s . d log k s . d

(1) 70 9 535 1 1 5 1.886 0. 13 1 3.7 33 0.3 28

NEMUBO (2) 7 0 3 533 1 25 1. 85 8 0. 1 37 3. 69 3 0.350

(3) 703 452 8 1 1.931 0. 1 26 3.7 1 3 0. 31 1

(4) 2215 1520 321 1.887 0. 1 30 4.185 0.3 27

(1) 67 2 451 65 1.864 0.0 55 3.52 1 0. 1 32

ASAHI- (21 662 450 74 1. 85 3 0. 05 3 3.5 05 0. 1 25

KAWA

(3) 630 397 53 1.86 5 0. 0 39 3.4 27 0. 09 9

(4) 1964 1 298 192 1.86 0 0.045 3.9 64 0. 1 04

(1) 1167 796 1 1 2 1992 0. 1 22 4.039 0.28 4 HAKODATE (2) I 162 796 1 1 2 1.96 7 0. 104 3.9 8 4 0.2 43

(3) 1062 472 62 2.00 1 0. 09 5 3.7 58 0.24 3

(4) 339 1 2064 286 1.984 0. 1 06 4.4 1 9 0. 25 0

(1) 1167 884 144 2.020 0.1 64 4. 1 58 0. 400

HACHINHE (2) 1 1 89 86 1 126 2.038 0. 164 4. 161 0.3 50

(3) 94 3 433 64 1.951 0. 089 3.66 0 0.2 23

(4) 3299 2178 334 2. 01 3 0. 1 4 1 4.5 1 6 0.34 2

(1) 1359 9 53 181 1. 800 0.0 47 3. 7 37 0. 120

SENDAI (2) 1349 954 169 1. 820 0. 04 6 3. 75 1 0. 1 22

(3) 1 293 588 85 1.920 0. 08 2 3.717 0.207

(4) 4001 2475 435 1.831 0. 0 50 4.19 5 0. 1 30

(1) 662 615 167 1.742 0.087 3. 53 9 0. 233

NI IGATA (2) 674 622 1 7 1 1.729 0. 0 80 3.519 0.21 0

(3) 575 446 94 1.8 6 6 0. 1 1 9 3.59 8 0. 3 06

(4) 1911 1683 432 1.762 0.087 4.0 04 0.230

(1) 646 536 69 2.08 2 0. 1 44 4.01 1 0.3 35

WAJIMA (2) 655 550 71 2.070 0. 1 49 4. 0 1 1 0.3 45

(3) 361 274 25 2.271 0. 22 1 4.03 6 0.494

(4) 1662 1360 165 2.105 0. 153 4.456 0.35 3

(1) 1219 983 186 1.868 0.09 2 3.9 07 0. 240

TOKYO (2) 1201 934 173 1.897 0. 1 09 3.94 9 0.27 4

(3) 922 646 1 1 7 1.9 1 9 0. 105 3.80 2 0. 26 6

(4) 3342 256 3 476 1.889 0. 100 4. 365 0. 2 56

Table 2 Samp \ ea of the cal cul ated values of m and log k and their standard

devi at ions (s . d) at various stations.

Nl. N2 and N3 respectively m ean

s

the total number, the number of data

of AS 10 and of A^2,0. Com ponent : i(l) N-S. (2) E-W. (3) U-D.

(4) resultant

III-7
to 20 10 10 10 3D

-J l_

-

' ' 10 ID 30

1 (a) Relations of N versus log A and m 'versus log Aj at Hikone station

«fos'i

*. ***.

' ' 10 10

1 (b) Relations of N versus log And m ^versus log Aj at Okayama station

III-8
id'

. 1 1

1

10 u

2 Relations of N versus log A and m versus log Aj at Tokyo station

o o

I

Ul.

NS • EW

UD <

ToUl-

•« ««

JO GO

SUIion

Values of m or log k obtained from three components and from resultant

many stations

III-9
III-IO

III-12
Fig. 8 (a) Regional distribution of the frequency when maximum displacement

amplitudes of larger than 100 micron will be observed in the comming

ICQ years

III-13
Fig. 8 (b) Regional distribution of the frequency when maximum displacement

amplitudes of larger than 1,000 micron will be observed in the comming

100 years

III- 14
Fig. 8 (c) Regional distribution of the frequency when maximum displacement

amplitudes of larger than 10,000 micron will be observed in the comming

100 years

III-15
Fig. 9(a) Regional distribution of maximum

displacement ampl i t udes ( x 10 cm)

which are expected to be observed

once at least in the comming 100 year s

Fig. 9 (c) Ebipectancy map of the

maximum acceleration

amplitudes (gal) in 100 years. Period of the d- ata used = 679 - 1948 (After H .Kawasum iCs})

Fig. 9 (b| Expectancy map of the maximum velocity am- plitudes (kine) in 50 years. Period of the data used = 1868 - 1964 (After I .Muramat3uC4])

Fig. 9 (d) Expectancy map of the maximum

velocity amplitudes of earthquake motions at bed rock in lOO years

.

(After K. Kanai et al.[63)

ii];-i6
joojoooooo^

"a

10 T

o Aomofi A Nogano

w

m

ZO

to -

01

• • • • AO o 8

30

log Aj

10 Relations of N versus log A and m versus log Aj (N -S component) at

Aomori, Akita and Nagano stations

IIIrl7
QUANTIFICATION OF SEISMICITY

by

Tsutomu Terashima and Tetsuo Santo

International Institute of Seismology

and Earthquake Engineering,

Building Research Institute

Ministry of Construction

ABSTRACT

A new seismicity index is proposed in this paper. This index is defined by the

following equation;

^N(M)

^T = ^ "t ^ A < 100 km ^ =

^

where, N(M) represents the number of shallow earthquakes with M > 6 which have occurred

within a epicentral distance A of 100 km during the period T(year). Using this criterion,

seismicity index maps in or near Japan for two different periods have been made.

Application of this index indicates that a remarkable change of S^ was found in

southern part of Boso Peninsula before and after Great Kanto Earthquake of 1923. Therefore,

in the field of earthquake engineering, these seismicity index maps can be used as a zoning

map of earthquake risk.

Key Words: Earthquake Distributions; Epicenters; Maps; Seismicity Index.

III-18
Introduction

Since C.F. Richter (1) defined the magnitude scale (M) of earthquakes, each earthquake

has been compared with each other quantitatively by this magnitude

.

Relative comparisons of seismicities among various regions, however, have never been

made quantitatively, therefore, an attempt to quantifying the seismicity is made in this

study.

Definition of Seismicity Scale

Fig. 1 shows epicentral distributions of 3147 remarkable and moderate earthquakes

which have occurred in or near Japan during a period from 1900 to 1950. In this figure, the

active areas are easily seen qualitatively. If we add the epicenters of smaller earthquakes

to this figure, however, the outline of the Japan Islands will disappear by the numbers of

dots of the epicenters. Therefore, by this ordinary method, it is very difficult to distinguish between active and nonactive areas

.

Fig. 2 shows another method of expressing seismicity by using different marks for

different magnitudes. This method, however, is also insufficient to distinguish quantatively the seismicity of various regions.

Consider the following seismicity index S^, which is defined here as a quantitatively

seismicity scale. *

where N(M) is the number of shallow earthquakes with magnitude greater than 6 which have

occurred within the epicentral distance of 100 km around a point. When T is taken as 100

years, we shall express S by removing the suffix T, and call the S as the "standard

seismicity index".

As seen in Fig. 3, world-wide distributed stations can detect the occurrences of all

earthquakes of M ^ 6. Therefore, the above definition aims to present only those earthquakes which have M >^ 6, which generally cause the most disaster. The limitation of A <^ 100

km in the definition of seismicity index S^ was made because the diameter of focal region

of the earthquake of M '=8 is almost 100 km in diameter. Tsviboi's [2] result also shows

that the diameter of the earthquake province in or near Japan is almost 250 km.

Seismicity Index S^ in or near Japan

A value of S equal to S at 7 35 points in or near Japan centering in the meshes with

T OD

0.5° of longitudes and latitudes was calculated during the period of 1885 through 1970.

As shown in Fig. 4, the total number of earthquakes measured were 1296, with the highest

area of seismicity index equal to 2. This means that earthquakes of intensites of M >^ 6

have occurred twice or more in a year. The area, with an index S„^ or 0.2 > S. - > 0.1,

86 — 86 =

therefore, represents those locations where earthquakes have the same magnitude which

have taken place once in 5 to 10 years. Seismicity regions of S ^ 2 is twenty times

86

larger than that of the region of S £ 0.1.

A < 100 km

M > 6

T

III-19
Tsuboi [3] has conducted similar studies as presented herein, however, since magnitudes

of earthquakes in or near Japan were not sufficiently defined at that time, areas of known

earthquakes were used instead of magnitudes. Tsuboi [3] index was called "Passive seismicity index number" as is shown in Fig. 5. The patterns of the contour lines on land as shown in Fig. 4 and Fig. 5 are very similar to each other, however the indexes are different as shown in Fig. 4, the mean frequency of earthquakes (M <^ 6) per year is seen clearly

but the mean frequency is not distinguished in Fig. 5.

A map of seismicity index 5^^(1926-1970) after the Kanto Great Earthquake (1923) is

shown in Fig. 6, which was made in order to see if the seismicity index changes before and

after that Great Earthquake. A remarkable difference is revealed in the southern Boso

Peninsula as shown in Fig. 4 and Fig. 6. In this area, the seismicity index is 2 >^ S-,->_ 1 — 85

as shown in Fig. 4, while the value is S^^ £ 0-1 shown in Fig. 6. This means that the

seismicity index has changed to a value of 1/10 of that of the total interval in this area. Dambara [4] has made a cumulative earthquake energy map in or near Japan as shown in Fig. 7.

An interesting fact in examining this data is that there is a close interrelation between

the contour lines in Fig. 6 and the darkness in Fig. It nas also been pointed out by

Mogi [5] and Brune [6] that low seismicity sometimes appears before the occurrence of a great earthquake. ,

Conclusions

A new method of estimating quantitiively the seismicity is proposed. In the field of

earthquake engineering, zoning maps for earthquake risk have been determined by a statistical method based on expected acceleration or intensity in various areas. The seismicity

index maps presented in the present paper have been developed without considering the

expected acceleration and or intensity in various areas . They have been made only to represent the seismicity in various areas precisely and quantitatively . The new seismicity

index map, however, will be available as zoning map for earthquake risk.

A seismicity index of S^^ after 1923 reduced to about 1/10 of Sg^ after 1885 in the

area of southern Boso peninsula, which is an area of high density earthquake energy.

The above result show that there is an interesting relationship between the occurrence of

large earthquakes and the change of S^.

In the future, similar seismicity maps will be developed for a certain magnitude

ranges and the work will be extended to microearthquakes

.

III-20
References

1. Richter, C.F. : An Instrumental Earthquake Magnitude Scale, Bull. Seism. Soc. Amer. , 25

(1935) , 1-32.

2. Tsuboi, C: Earthquake Province - Domain of Sympathetic Seismic Activities Jour. Phys.

Earthq., 6(1958), 35-49.

3. Tsuboi, C: Passive Seismicity Index Number in Japan and Vicinity, Bull. Geophys. Inst.

Univ. of Tokyo, 12(1947), 1-6.

4. Damabara, T. : Accumulation of Earthquake Energy (in Japanese) , Rep. Coord. Com.

Earthq. Prediction 5(1971), 74-78.

5. Mogi, K. : Some Features of Recent Seismic Activity in and Near Japan (1). Bull.

Earthq. Res. Inst., 46(1968(, 1225-1236.

6. Brune, J.N. and C.R. Allen: A Microearthquake Survey of the San Andreas Fault System

in Southern California, Bull. Seism. Soc. Amer., 57 (1967(, 277-296.

III-51
I2flr 130 132 134 136 138 UO U2 \UU U6

130' 132 13A 136 138 UO U2 U/i

Fig. 1 Epicentres of 3147 earthquakes which occurred in or near Japan in 1900 - 1950 (Felt beyond 200 km)

III-22
Fig. 2 Epicentral distribution of shallow earthquakes

which occurred in or near Japan in 1885 ~ 1970.

(M |6.0)

III-23
III-24
4 Seismicity map of index Sg^ ( 1885 - 1970 )

.

III-25
Fig. 5 The map of "Passive seismicity index number"

(After C. Tsuboi [3] ),

III-26
Fig. 6 Seismicity index map of S.c (1926 ~ 1970),

III-27
Fig. 7 Accumulation of earthquake energy in or near Japan

(After Dambara [4] ).

III-28
MAINTENANCE OF THE STRONG-MOTION ACCELEROGRAPH

AND THE DATA PROCESSING OF THE RECORDS

by

Eiichi Kuribayashi

Chief, Earthquake Resistant Structures

Section, Structure and Bridge Division,

Public Works Research Institute, Ministry

of Construction

and

Hajime Tsuchida

Chief, Earthquake Resistant Structures

Laboratory, Structures Division, Port

and Harbour Research Institute, Ministry

of Transport

and

Makoto Watabe

Chief, International Institute of Seismology and Earthquake Engineering, Building

Research Institute , Ministry of Construction

ABSTRACT

The observation of strong-motion earthquakes, located at harbour , public works and

buildings throughout Japan by the national research institute has been made. These observations will be discussed, herein, in addition to such items as the maintenance and checks

of the strong-motion accelerographs , main records, processing and analyses of records, and

the availability of the data to the public.

Key Words: Accelerographs; Earthquake Data; Earthquake Records; Field Observations;

Strong-motion Accelerographs.

IV1
Introduction

1. History of strong-motion earthquake observations in Japan

The SMAC type strong-motion accelerograph now being used in Japan and in Central and

South America, Southeast Asia, and Middle and Near East, was developed by the Strong-Motion

Earthquake Observation Committee. The development of this instrument required 2 years and

was supported by science research funds of the Ministry of Education in 1950. The instrument has been improved since then and is more suitable to Japanese climate in the resistance

against moisture, cold, etc.

In October, 1953 after completion of the strong-motion accelerograph, the Resources

Council, Prime Minister's Office recommended the following to the prime minister; "to secure

the source of necessary revenue and to improve the setup and consolidation of the necessary

strong-motion earthquake observation network and to establish a system required for its

operation, in approximately 3 years," and further recommended in January, 1955 that permanent station plan of strong-motion accelerographs be established.

Later, in view of the disaster caused by the Niigata Earthquake i- 1964, the Science

Council of Japan recommended to the prime minister in November, of that same year; execute

a concentrated set of stations to measure strong-motion earthquakes". This recommendation

asked for stations at every 50 km throughout the country, thus providing one yardstick for

observation density. Further in October, 1971, the Science Council of Japan requested the

prime minister to examine the previous recommendations, given earlier.

From 1953 to the end of March, 1974, about 850 strong-motion accelerographs were installed. When the observations were started, the communication among the observation agency

and the publicity service were performed by the Strong-Motion Earthquake Observation

Committee, which is controlled by the Investigation Board of Resources, Prime Minister's

Office. In December, 1955, the controlling authority of the Committee was transferred to

Earthquake Research Institute, University of Tokyo. In 1960, the principal strong-motion

earthquake records were published by the Committee.

The number of SMAC strong-motion accelerographs, installed after the Niigata Earthquake^

increased suddenly in comparison with previous installations, and thus it is difficult for

the Strong-Motion Earthquake Observation Committee to examine these. In this situation, the

Liaison Conference for Promotion of Strong Motion Earthquake Observation Project comprising

concerned government offices, heads of concerned agencies and specialists, was established in

the National Research Center for Disaster Prevention, Science and Technology Agency. The

conference group decided to be responsible for communication and coordination of the nationwide promotion of the strong-motion earthquake observation project. The conference participants analyzed the present state of stationary strong-motion accelerographs , and discussed

how to increase the observation network, and at the same time provided a publicity service

for collection of the principal records. In February, 1972, the group arranged and published

their opinions regarding the improvement of the observation network. According to the

publications, they proposed to establish 2,141 strong-motion accelerograph

stations throughout the country. Therefore, even if all the strong-motion accelerographs

IV2
presently installed are considered to be effectively operating, installation of about 1,300

additional strong-motion accelerographs is necessary.

2. Necessity to strengthen and enlarge the strong-motion earthquake observation

network.

The strong motion earthquake observations stated herein, refers to the measurement of

acceleration amplitude waveforms caused by tremors of the ground and structures at the time

of a strong-motion earthquake. Such measurements relate to the reading of the measured

waveforms, and to the primary analyses of the magnitude of the acceleration amplitude,

periodical characteristic of the waveforms, etc.

Strong-motion earthquake observations and the subsequent

analyses of the magnitude of the acceleration amplitude and the periodical characteristic

of the waveforms has resulted in relationships between

the magnitude of the earthquake, epicenter distance, seismic center depth, topography,

geology, scales, construction, and material of the structures.

The analytical results of the strong-motion earthquake observations give quantitative

information and thus will permit an establishment of countermeasures for the earthquake

disaster prevention for each region of the country and for type of soil. It further

rationalizes the seismic design of structures built on the ground, and moreover rationalizes the aseismatic design of the structures which grow diverse, massive, and complex,

considerably contributing to safe and economical constructions of these structures. An

example of such a highly dense network, for measuring strong-motion earthquakes, has been

established around Los Angeles, California, USA. These observations were initiated in 1932,

and have since obtained 250 strong-motion accelerographs records prior to the 1971 Los

Angeles Earthquake.

In this country, strong-motion earthquake observations were initiated in March, 1953,

when a strong-motion accelerograph (SMAC type) made in Japan, was installed in the Earthquake Research Institute, University of Tokyo. Since then, various universities, government agencies, and various private groups have cooperated in the enlargement and strengthening of the strong-motion earthquake observation network. Finally at the end of March, 1974,

the total number of SMAC strong-motion accelerographs stationed throughout Japan was about

850, and though not reaching the required density.

The strengthening and enlargement of the strong-motion earthquake observation network

is necessary in order to achieve the aim described previously, thus it is still desirable

to increase the density of this still sparse observation network.

Strong-Motion Earthquake Observations in Ports and Harbours

1. Present state of the observation network

The strong-motion earthquake observations of port and harbour areas in Japan, have been

executed since 1962 under the cooperation of Port and Harbour Research Institute and various

agencies concerned with ports and harbours. This has been executed under an agreement

initiated by a research group consisting of the Port and Harbour Bureau of Ministry of

Transport, respective District Port Construction Bureaus, Port and Harbour Research Institute, and other concerned agencies. As of December, 1974, the observation network consists

IV3
of 65 strong-motion accelerographs , installed at 44 ports. Figure 2-1 shows the locations

of the ports and harbours, where the strong-motion accelerographs are installed. Figure

2-2 shows the number of installed accelerographs per year. There are 2 types of strongmotion accelerographs that used in observation network: one was a SMAC-B^ strong-motion

accelerograph and the other a ERS strong-motion accelerograph. The description of the

SMAC-B has been reported previously (2-1) . The ERC strong-motion accelerograph is an accelerograph with a moving type coil transducer, which is available as Type A with a magnetic tape recorder, and as Tape B or C with an electromagnetic oscillicgraph . Types A and

B record 2 horizontal components, and Type C records 2 horizontal components and a vertical

component. The ERS strong-motion accelerograph can be used with the transducer and the

recorder. This provides an advantage, for observing the earthquake response of port and

harbour structures in which strong-motion accelerograph rooms are impossible to make.

2. Maintenance and check of strong-motion accelerographs

The strong-motion accelerographs, installed in port and harbour areas are maintained

by the construction work offices, etc. at the respective ports. The staff in charge of the

maintenance and check of the strong-motion accelerographs (normally the staff for public

works) are assigned to conduct periodical checks (twice a month) and to perform special

checks after the occurrence of earthquakes of II or more in the earthquake intensity scale

of Meteorological Agency.

Also, once or twice a year, a detailed check is made by the manufacturer engineers of

the strong-motion accelerographs. In addition, the Port and Harbour Research Institute conducts training of the staff in charge of operations, every year. The training is practical,

and relates to handling of the strong-motion accelerographs by the engineers dispatched

from the manufacturers of the strong-motion accelerographs. A strong-motion accelerograph

is initiated when the starter senses the tremors of magnitude 50 to 10 gals. When a strong

motion accelerograph does not start at the time of an earthquake, it is difficult to judge

whether the tremors were too small or there was failure of the accelerograph. For this

reason, a strict acquisition rate of records cannot be shown. However, what will be introduced is a case to show the record acquisition state of the observation network. At the

time of the Tokachi-Oki Earthquake in 1968, 15 strong-motion accelerographs located in the

ports and harbours of Hokkaido, and Tohoku district recorded the tremors. Only one accelerograph did not record, due to its own failure.

3. Principal records

From those records obtained by the strong-motion earthquake observations in port and

harbour areas, those exceeding about 100 gals in the maximum acceleration are shown in

Table 2-1 and are classified by each observation point. As an example of recorded waveforms, the record of the Tokachi-Oki Earthquake in 1968 at Hachinohe Port is shown in Fig.

2-3,

4. Processing of records

a) Storage of records

All the records obtained at the respective observation points, are sent to the

Port and Harbour Research Institute, to be catalogued and stored. Copies of

IV4
records are also sent to the respective observation points . The records ac- quired are classified for each earthquake every 2 months, and with a "StrongMotion Earthquake Observation Table" showing the maximum acceleration of each

component of the respective records, which are sent to the parties concerned,

b) Digitization of records

For those records which were obtained on the ground and exceed 50 gals in maxi- mum acceleration, the horizontal components are digitized.

The record by an SMAC-B^ strong-motion accelerograph is copied by contact printing on Mylar-based sensitive paper, and is digitized by a semi-automatic digitizer. The digitizer

places a cursor on the position requiring digitization, coordinate values are then punched

on a paper tape by pressing a switch. The time intervals used during the digitization, are

0.01 sec. The numerical values which were punched on the paper tape are then read by an

electronic computer, and reproduced as waveforms by a graphic output unit in order to con- firm that the readings were correct. Any errors noticed are then corrected. The corrected

records provide amplitude values at unequal time intervals , and are converted into amplitude

values at the equal time intervals of 0.01 sec by interpolation, and are then recorded on magnetic tape. In principle, the zero-line is not corrected, however when the zero-line is

not properly inserted for those waveforms reproduced by the graphic output unit, the

digitization is re-evaluated. However, if time is not available because of the many records

to be processed, the waveforms are enlarged by the graphic output unit, in order to determine the corrective amount ot the zero-line position, for correction in the electronic coifl- puter

.

The records from the ERS-B/C strong-motion accelerograph is digitized using a different

digitizer than that used for the SMAC-B^ strong-motion accelerograph. This digitizer is

interlocked with a computer and if an operator moves the cursor along the recorded waveforms*

an amplitude value is stored in the computer each time the cursor moves 0.1 mm along the

time base.

After end of digitization, the numerical values stored in the computer are reproduced

by a digital-to-analogue converter as an analogue voltage which are recorded by a penwriting oscillograph. After conformation that the recorded waveforms have been correctly

digitized, numerical values are punched to the paper tape. They are then read by another

computer, and recorded on magnetic tape together with the records of the SMAC-B strongmotion accelerographs

.

No corrections are made, to the ERC-B/C strong-motion accelerograph records. The

minimum time intervals for digitization, relate to the recording paper feed rates of the

respective strong-motion accelerographs. These rates are 0.005 sec. for Type B and 0.0025

sec. for Type C. However, the records which are stored on the magnetic tape, have the same

time intervals as those of SMAC-B strong-motion accelerograph. It was experimentally confirmed, that the digitization of the ERC-B/C strong-motion accelerograph records, could be

reproduced with an accuracy which causes no viewing difficulty for practical use.

IV5
c) Spectral analysis

For all the digitized records, the response spectra and Fourier spectra are calculated. Other various spectra are calculated according to necessity.

5. Opening of data

The records obtained by the strong-motion earthquake observations in port and harbour

areas, are published in the "Annual Report on Strong Motion Earthquake Records in Japanese

2-2)

Ports" . The annual report covers the maximum acceleration of each component of every

record, the seismic center location, magnitude, seismic intensity at the respective location

for the earthquake corresponding to each record. Moreover, the ground recordings for those

with accelerations of 20 gals or greater have their waveforms reproduced, and for those

with maximum accelerations of 50 gals or greater have their waveforms digitized with response spectra and Fourier spectra provided.

2-2)

The published annual reports extended from 1963 to 1974 . Also the record of the

2-3) Tokachi-Oki Earthquake in 1968, has been published separately

The data where each strong-motion accelerograph has been installed, are published as

"Site Characteristics of Strong Motion Earthquake Stations in Ports and Harbours in Japan"

2-4 2-5 2-5)

' ' . The data at these observation points, covers the geographical features

where the installations are located such as the configurations of the buildings, foundations

of the strong-motion accelerographs , design drawings of the strong-motion earthquake observation rooms, and soil histograms. The results of these observations are given in the

annual report as in the analytical results such as average response spectra and seismic

2—7 2—8)

tremor characteristics of the ground '

Strong-Motion Earthquake Observations at the Public Works Research Institute

1. History of strong-motion earthquake observations at the Public Works Research

Institute

The strong-motion earthquake observations at the Public Works Research Institute, were

initiated in 1957 when an SMAC type Strong-motion accelerograph and an electromagnetic

strong-motion accelerograph were installed in Kinki Regional Construction Bureau. Seventeen

years later, at the end of March, 1974 a total of 189 strong-motion accelerographs were installed for the public works facilities using the research funds of the Public Works Research

Institute and funds from various civil engineering construction projects.

This report describes the history, system and future promotion of strong-motion earthquake observations at the Public Works Research Institute, and refers to the entire nation

and Ministry of Construction projects and their inter-relationship.

2. Progress of the strong-motion earthquake observations

a) Observation stations

The strong-motion earthquake observations , at the civil engineering work facilities such as bridges, dams, embankments and tunnels are executed under the

cooperation of various agencies. These agencies consist of the Public Works

Institute, Regional Construction Bureaus, local governments (Civil Engineering

Works Department, Board of Project) , 4 public corporations concerned with

IV6
civil engineering works, and the Hokkaido Development Agency. Since the observations were started in 1957, with the use of an electromagnetic type strongmotion accelerographs at the Sarutani Dam in the Kinki Regional Construction

Bureau, additional observation stations have increased gradually. Since the

Niigata Earthquake in 1964, the installation of additional strong-motion

accelerographs was promoted and increased by notification of the Director of

the Road Bureau. At the end of March, 1974, 189 SMAC type strong-motion

accelerographs were installed at 96 stations and electromagnetic type strongmotion accelerographs with 448 components, were installed at 48 stations. The

locations of the observation stations of the SMAC type strong-motion accelerographs and the electromagnetic type strong motion accelerographs are shown in

Figs. 3-1 and 3-2. The number of installed accelerographs per year, are shown in Table 3-1, and the numbers of installed accelerographs per structural

type are shown in Table 3-2.

b) Strong-motion accelerographs

The strong-motion accelerographs used for observation are the SMAC type

strong-motion accelerographs and electromagnetic type strong-motion accelerographs, with the SMAC type, especially Type B^, used predominantly. However,

recently. Type and Type Q, which allows one to measure the components in a shorter period than permitted by Type B^, were installed. The electromagnetic

type strong-motion accelerograph uses an electromagnetic method for the transducer, and an electromagnetic oscillograph or magnetic recorder for the recorder. In using a transducer for the strong-motion accelerograph, the moving

coil type accelerometer is used most often, variations of the transducer in

size and weight, is considered dependent on the installation location and where

the SMAC type strong-motion accelerograph cannot be installed. Recently,

observations of seismic tremors in the ground under the surface of the ground

have been executed. At present such observations, by the electromagnetic type

accelerographs, have been made at 15 stations (149 components) out of 48

stations . The number of measuring components and units used of each electromagnetic type strong-motion accelerograph differ according to each station.

c) Observed records

The records obtained from the initiation of the observations until the present

consist of about 3,817 Sheets (including 869 Sheets for the Matsushiro Earthquake Swarm) as recorded by the SMAC type strong-motion accelerographs , and

about 210 Sheets by the electromagnetic type strong-motion accelerographs.

Most of the records listed values of less than 100 gals, and the number of

records of 100 gals or more were about 77 (including about 30 records for the

Matsushiro Earthqviake Swarm) as shown in Table 3-3. The largest recorded

waveforms obtained to date were from Uwajima City, Ehime Pref . , which had a

maximum acceleration of 438 gals, with the waveforms as shown in Figs. 3-3 and

3-4.

IV7
3. Observation net-work

The strong-motion accelerographs was installed according to the public works con- struction sites. Since the past several years, the increase in the number of installed

accelerographs has shown a tendency toward an uneven distribution in specific areas, this

then shows what areas need accelerographs installed. The ideal plan, for stationing

strong-motion accelerographs, is to obtain a record in the adjacent area of the epicenter

when an earthquake with V or more seismic intensity (JMA) occurs in any part of Japan. The

present status, however, makes it impossible to approach these ideal conditions in the near future. Therefore, the following network has been drafted to promote a strong-motion

earthquake observation project.

The network involves the installation during the next five years (1975-1979) , and to

install others according to necessity. The network plan was drafted per the following

conditions

:

(1) The entire land area of Japan should be covered with 258 equilateral triangle

meshes (one side is about 50 km) , for the stationing at least 1 observation station in

1 mesh.

(2) Covering all the meshes in Japan, uniformly with extreme urgency.

(3) At least one each, at the top and bottom of a structure, with a total of two

sets, should be installed at one observation station.

(4) Consideration of those installed strong-motion accelerographs , from other ministries, should be taken into consideration.

In order to determine the total number of strong-motion accelerographs to be installed

in the future, the present stationing state was investigated as shown in Fig. 3-5. The

meshes, without strong-motion accelerographs can be selected as dark meshes shown in Fig.

3-6, and where, 146 stations (292 sets) require installation. The complete installation is

planned for completion in the latter 3 years of the five-year project. The draft of the

network is still under adjustment with the sponsoring organizations.

4. Checks and collection of records

The maintenance and control of the strong-motion accelerographs, related to the

civil engineering works facilities, have been executed by the notification from the Director

of the Roads Bureau and the Director of the River Bureau These notifications were deliveres

in April, 1970, and have been instituted by the offices controlling the facilities in

which the strong-motion accelerographs were installed. Examination of the strong-motion

accelerographs involve periodic checks and special checks by the staffs of respective

offices. Also technical checks by special engineers, of the manufacturers of the strongmotion accelerographs, have been made.

a) Periodic check

A check of the equipment has been conducted every 2 weeks, in order to confirm

the preparedness of the equipment

b) Special check

When there is an earthquake of II or more on the seismic intensity of JMA, in

the neighborhood of the observation station , a special sheck is

IV8
conducted to collect a record,

c) Technical check

This check is conducted, by a special engineer of the manufacturer of the strongmotion accelerograph , once or twice a year.

5. Collection, storage, analyses and publication of the records

All the seismic records obtained at the respective observation stations are mailed to

the Public Works Research Institute for arrangement and storage. The arrangment and

analyzing work of the seismic data and records are performed based on the flowchart as

shown in Fig. 3-7. When the records are obtained, those with a maximiam acceleration of .50 gals

or more are digitized. The digitization is performed by use of a semi-automatic digitizer,

according to the flowchart as shwon in Fig. 3-8. The resolution of the digitizer is 25

microns, and the overall precision is 0.125 mm. The digitized records are subjected to a circular correction (in case of SMAC-B^) , a zero-line correction and a time base correction

by an electronic computer. The data is converted into numerical values at 0.01 sec. intervals, by linear interpolation. Also developed from the waveform analyses, are autocorrelation coefficients, Fourier spectra, power spectra and various response spectra which are digitized and is further processed for graphic output unit.

The pioblication of the observation data and the strong-motion earthquake records,

are piiblished every year. The recorded waveforms of the previous year are copied in fullscale, and arranged in order to give of the various contrast seismic values. The tables of

the numerical values are also published, when a predetermined voliame of data is accumulated.

The data published to date are contained in 12 books (references 3-1 to 3-6, 3-9, 3-11, 3-13,

3-16, 3-17) for strong-motion earthquake records and 5 books (3-7, 3-8, 3-14 and 3-15) for

tabular nianerical values

.

Strong-motion Earthquake Observations at the Building Research Institute

1. History of strong-motion earthquake observations, concerned with building

In 1954, a strong-motion accelerograph Type SMAC-A, was installed at the Building

Research Institute by the Director of Housing Bureau, Ministry of Construction. The observation network, close to the number of present strong-motion accelerographs , was then

formed in the 1960s. The types of strong-motion accelerographs that were installed, were

Type SMAC-A, Type DC developed jointly with the Research Institute, and Types SMAC-B^ and

SMAC-B2, improved version of SMAC-A. Since the initially installed SMAC-A type became

superseded, the SMAC-M type was intended for complete automation and digitization by the

2nd joint development with the Research Institute. This type has been used since 1973 as a

strong-motion accelerograph. In 1974, the mobile observation network, comprising several

strong motion accelerographs commonly called "Caravan" was initiated for installation.

This is to be preferably installed for a period of several years in the place most likely

to be subjected to strong-motion earthquakes according to the information relating to the

prediction of earthquakes. Also, if the intended strong-motion record can be obtained, it

is to be moved to the next place with high earthquake occurrence probability. In the

building field, the Earthquake Research Institute of the University of Tokyo has been

performing strong-motion earthquake observations together with this research institute.

IV-f
However, in order to maintain the increased number of installed strong-motion accelerographs

the National Research Center for Disaster Prevention and the Science and Technology Agency

have shared a part of the observation network. The strong-motion accelerographs related

to buildings are primarily installed in the multistory structures with 45 m or more height these buildings are constructed by special permission of the Minister of Construction according to the stipulations of Article 38 of the Building Standard Law. The number

of installed strong-motion accelerographs of this kind has sharply increased in accordance

with the increase of construction of multistory buildings as shown in Fig. 4-1. However,

the maintenance and control capacity of the Building Research Institute Earthquake Research

Institute of University of Tokyo, the National Research Center for Disaster Prevention, and

parts of private and public corporation, are beyond their capacity.

2. Present state of strong-motion earthquake observation network related to buildings

The Building Research Institute installs strong-motion accelerographs about every

100 km along the coasts throughout Japan. These installations are mainly in the neighborhood of the foundations of the buildings. As shown in Fig. 4-2, 31 strong-motion accelerographs are installed in 18 places. There are 8 sets of Type SMAC-A, 6 sets of Type SMAC-B

9 sets of Type SMAC-M and 8 sets of Type DC.

The numbers of strong-motion accelerographs relating to buildings and the agencies in

charge of maintenance and control are shown in Table 4-1. In addition, the Building

Research Institute set up the afore-said "Caravan" at 2 places. Each Caravan has strongmotion accelerographs on the foundations in the neighbothood of the vertexes of an equilateral triangle with about 1 km sides. Recorded are the absolute time and common time in

order to allow accurate acquisition of the propagation direction of seismic waves, etc.

One Caravan is installed in the Numuro Penninsula, which is said to have residual energy

in the seismic center, with the expected seismic scale similar to that of the previous time,

and the other Caravan is installed in the neighborhood of "Omaezaki" facing the Sea of

Enshu since the next big earthquake is expected to occur in the Sea of Enshu trough with a magnitude of 8.

3. Maintenance and check of strong-motion accelerographs related to building

The organization system of maintenance and control in the Building Research Institute

is shown in Fig. 4-3. The Building Research Institute asks local engineers concerned with

the buildings at the observation points to perform maintenance and check at the observation spots with due renumeration. The local staffs performed a monthly check, and when

they find any trouble and shortage of spare parts, etc. hey inform the Building Research

Institute. The Building Research Institute then dispatches a repairman or sends spare

parts to the location. When a strong-motion earthquake has occurred in the neighborhood

of any location, the local staff instantly checks the strong-motion accelerograph, and if

there is any record available, he enters the data and puts the record in a metallic

cylinder. The cylinder is then sent by amil to the Building Research Institute, irrespective of the magnitude of the record. Moreover, in order to improve the technique

knowledge of the naintenance and checks performed by the staff concerned with strong-motion

earthquake observations, a short study course has been given every three years by

IV- 10
Building Research Institute. The maintenance and check of the strong-motion accelerographs

related to buildings are vary in their systems and organizations, etc. In agencies other

than the Building Research Institute consist of the following:

An agency in charge of observations has one responsible technician concerned with,

strong-motion earthquake observations and he maintains, checks,

ment.

The actual maintenance, checks and repairs are performed by an external engineering

contractor at an interval once or twice a year. The strong-motion earthquake records

are collected by the responsible technician of the agency or the external engineering

contractor. A common trouble with all the agencies is obtaining the budget for

maintenance, checks, and repairs. This is because the fund must be raised from the

budget under another item, and does not appear as an independent budget.

4. Processing method of strong-motion earthquake records

Since the record processing method is about the same among all agencies related to

buildings, the processing method employed by the Building Research Institute will therefore

be introduced herein. The Building Research Institute uses roughly 3 types of strongmotion accelerographs: (1) those obtained by smoked paper recording, (2) those obtained

by stylus paper recording, and (3) those obtained by magnetic tape recording. In the

former two types, the original record is copied by contact with the negative plate, and is

digitized according to the flowchart shown in Fig. 4-2, the data is then analyzed by the

computer, for generation of a list of digitized data. In case of the magnetic tape, the

record is directly digitized by a high speed analogue- to-digital converter, and then

analyzed as described in Fig. 4-2. The utilization of strong-motion earthquake records

do not require immediate evaluation. However, the publishing of the analyzed results,

before general engineers and administrators lose their interest in the earthquake, is

" necessary in order to provide them with the significance of the strong-motion earthquake

observations. Therefore, the Building Research Institute has prepared a completely new

process for all records, thus permitting the data to be distributed within 4 days after

collection.

5. Publication of records

The strong-motion earthquake observation records related to building, which have

acceleration values exceeding 40 gals , are sent to the Liaison Committee for Promotion

of the Strong-Motion Observation Project. These records are then published semi-annually

as original records (analogue values) including the additional information relative to the

seismic center, seismic scale, time magnitude, and seismic scale distribution at various

places. The Building Research Institute published the digitized tables and analized results

; of the main strong-motion earthquake records in Annual Research Work of the Building Rej. search Institute, the B.R.I. Research Paper, the News of Japan Society for Earthquake

I Engineering Promotion, etc. In addition, about once a decade, the "Digitized Strong-Motion

Earthquake Accelerogram in Japan" is also published. This data contains earthquakes location of each observation point, incessant tremor Fourier spectra of the point, the external

view photo and sectional view of the building in which each strong-motion accelerograph is

IV-11
installed, strong-motion records and their digitized tables, response spectra of dislocation, speed and accelerations.

6. Examples of the principal strong-motion earthquake records

Table 4-2 shows the records which have a meiximum acceleration exceeding 100 gals, and

from those records obtained on the 1st floor of foundation of the building. Also from these

records illustrated are the data of the strong-motion earthqucike record obtained at

Kawagishi-cho Apartment of the occasion of the Niigata Earthquake (June 16, 1964).

This earthquake was the most significant in the history of strong-motion earthquake observations in Japan, and the record of the strong-motion accelerograph of Hiroo Town Office

on the occasion of the Hitaka Sandei Earthquake (July 21, 1970) which recorded the largest

acceleration as of this date.

IV-tl2
References

(2-1) Present Status of the Observations of Strong-Motion Earthquakes in Japan, T. Ariga,

The 1st Joint Meeting of Wind and Seismic Effects, U.J.N.R. , 1969.4

(2-2) Annual Report on Strong-Motion Earthquake Records in Japanese Ports (1974), E. Kurata

I. Ishizuka and H. Tsuchida Technical Notes of the Port and Harbour Research

Institute, No. 202, 1975.3

(2-3) Strong-Motion Earthquake Records on The 1968 Tokachi-Oki Earthquake and its After

Shock, H. Tsuchida, E. Kurata and K. Sudo, Technical Notes of the Port and Harbour

Research Institute. No. 80, 1969.6

(2-4) Site Characteristics of Strong-Motion Earthquake Stations in Ports and Harbours in

Japan (Part I) , H. Tsuchida, T. Yamada and E. Durata Technical Notes of the Port

and Harbour Research Institute. No. 34, 1967.11.

(2-5) Site Characteristics of Strong-Motion Earthquake Stations in Ports and Harbours in

Japan (Part II) , E. Kurata, H. Tsuchida and K. Sudo Technical Notes of the Port and

Harbour Research Institute. No. 107, 1970.12

(2-6) Site Characteristics of Strong-Motion Earthquake Stations in Ports and Harbours in

Japan (Part III) , E. Kurata and I. Ishizuka, Technical Notes of the Port and Harbour

Research Institute. No. 156, 1973.3

(2-7) Average Response Spectra for Various Subsoil Conditions, S. Hayashi, H. Tsuchida,

and E. Kurata, The 3rd Joint Meeting of Wind and Seismic Effects, U.J.N.R. 1971.5

(2-8) Characteristics of Baserock Motions Calculated from Strong-Motion Accelerograms,

S. Hayashi, H. Tsuchida and T. Uwabe, The 5th Joint Meeting of Wind and Seismic

Effects, U.J.N.R. 1973.5

(3-1) PWRI, "Strong-Motion Earthquake Records from Public Works in Japan (The Matsushiro

Quake Swarn, December 9, 1965 - February 28, 1966, The Ochiai Bridge)", Technical

Memorandum of P.W.R.I., No. 163, April 1966.

(3-2) PWRI, "Strong-Motion Earthquake, Records from Public Works in Japan (The Matsushiro

Quake Swarn, March 1 - June 22, 1966, The Ochiai Bridge)", Technical Memorandum of

P.W.R.I., No. 194, July 1966.

(3-3) PWRI, "Strong-Motion Earthquake Records from Public Works in Japan (1961-1966)",

Technical Memorandum of P.W.R.I., No. 217, December 1956.

(3-4) PWRI, "Strong-Motion Earthquake Records from Public Works in Japan (The Matsushiro

Quake Swarn, June 25-December 1, 1966, The Ochiai Bridge)", Technical Memorandum of

P.W.R.I., No. 254, March 1967.

(3-5) PWRI, "Strong-Motion Earthquake Records from Public Works in Japan (The Matsushiro

Quake Swarn, Mary 27-October 23, 1966, The Susobana Dam)", Technical Memorandum of

P.W.R.I., No. 263, March 1967.

(3-6) PWRI, "Strong-Motion Earthquake Records from Public Works in Japan (1967)", Technical

Memorandum of P.W.R.I., No. 341, February 1968.

(3-7) PWRI, "Digitized Data of Strong Motion Earthquake Records (No. 1)", Technical Memorandxam of P.W.R.I., No. 317, March 1968.

(3-8) PWRI, "Digitized Data of Strong-Motion Earthquake Records (No. 2)", Technical Memorand

um of P.W.R.I., No. 318, March 1968.

(3-9) PWRI, "Strong-Motion Earthquake Records from Public Works in Japan (1968)", Technical

Memorandum of P.W.R.I., No. 430, April 1969.

IV-IS
(3-10) PWRI, "Digitized Data of Strong-Motion Earthquake Records (No. 3)", Technical Memorandum of P. W.R.I. No. 461, May 1969.

(3-11) PWRI , "Strong-Motion Earthquake Records from Public Works in Japan (1969)", Technical

Memorandum of P.W.R.I. No. 641, March 1971.

(3-12) PWRI, "Strong-Motion Earthquake Records from Public Works in Japan (1970)", Technical Memorandum of P.W.R.I. No. 718, March 1972.

(3-13) PWRI, "Strong-Motion Earthquake Records from Public Works in Japan (1971)", Technical Memorandum of P.W.R.I., No. 805 March 1973.

(3-14) PWRI, "Digitized Data of Strong-Motion Earthquake Records No. 4", Technical Memorandum of P.W.R.I. No. 877, December 1973.

(3-15) PWRI, "Digitized Data of Strong-Motion Earthquake Records, Revision (No . 1-No. 3) "

,

Technical Memorandum of P.W.R.I. No. 876, December 1973.

(3-16) PWRI, "Strong-Motion Earthquake Records from Public Works in Japan (1972)", Technical

Memorandum of PWRI No. 713, March 1974.

(3-17) PWRI, "Strong-Motion Earthquake Records from Public Works in Japan (1973)", Technical

Memorandum of P.W.R.I. No. 967, November 1974.

IV-14
iy-15
IV-16
IV- 17
Fig. 3-1 Sites of Strong-Motion Seismograph Stations for SMAC-Type

Accelerographs. - Observation Network of Ground Motions

and Earthquake Responses of Highway Bridges, Tunnels,

Dams and Embankments - as of March, 1974

IV-18
Fig. 3-2 Sites of Strong-Motion Seismograph Stations for ElectroMagnetic Type Seismographs. - Observation network of

Ground motions and Earthquake responses of Highway bridges,

Dams and Embankments - as of March, 1974

IV- 19
r 1

200 r

100

0

100

200

Tronflvtrse

Itajima Bridge 308-P-4

1968 8.6 01:17 The BungoBLido Earthquake

200

100

100

200

200

100

Longitudinal

0 ^a*AVAH.VVV^ li MlI ?l

Fig. 3-3 Acceleration records at the Itajima Bridge during the

Bungosuido Earthquake of August 6, 1968

100

200

Itajima Bridge 308-Gr-*

1968 8.6 01:17 The Bungosuido Earthguake

200

lOO

100

200

gal K4m

Longitudinal

100

200

Fig. 3-1 Acceleration records on the ground surface nearby the

Itajima bridge during the Bungosuido Earthquake of

August 6, 1 968

IV- 20
Fig. 3-5 Distribution of Strong-Motion Observation Stations of Japan,

as of March, 1974

IV- 21
Fig. 3-6 Proposed networks of Strong-Motion Observation Stations

of Ministry of Construction, as of March, 1974

• installed, P. W.R.I.

)( installed, B.R.I,

A installation expected in 1977 F.Y., 48 stations

A installation expected in 1 978 F .Y . , 50 stations

A installation expected in 1979 F.Y., 40 stations;

O Number with circle means expected number of iL

stations in isolated islands

O

IV- 2 2
Site

Conditions

Seismic

Conditions i

Collection

of StrongMotion

Records

Read

Maximum

Acceleration

Digitize

Records

Card Card Card

Computer

Processing

Conditions

for

Processing

Conditions

j for Processing

MT 3

Table of Properties

of Observed

Earthquakes

Table of Records

for Each

Earthquake

Table of Conditions of Observation

Stations

Table of Records for

Each

Statiori,

Plotting

Microfilm

Digitized

Data

Plotting

Microfilm

Fig. 3-7 Flowchart for Processing Strong-Motion Observation Data

(Strong-Motion Records and Seismic Conditions)

Analysis

Plotting

Microfilm

IV-2 3
IV- 2 4
IV- 2 5
IV- 2 6
Fig. 4-3

SYSTEMS FOR STRONG MOTION OBSERVATION IN B.R.I.

a) DECISION MAKING

CHAIRMAN

RESEARCHER

FOR SYSTEM

ACTING-CHAIRMAN

HEAD OF DIVISION

CONCERNED

HEAD OP

PLANNING

CHIEF OF BUDGET

SECTION

b) MAINTENANCE AND DATA PROCESSING

ACTING-CHAIRMAN

MAILING

OBTAINED

DATA

RESEARCHER

MONTHLY

REPORT

SUPPLY OF

DRY CELL

E.T.C.

ENGINEER AT

THE SITE

X

MONTHLY

CHECK

STRONG MOTION

ACCELEROGRAPH

DATA PROCESSING

PUBLISHING

TECHNICIANS FOR

MAINTENANCE

FIXING

SETTING

OF THE

INSTRUMENT

IV- 2 7
IV- 2 8
Accelerogram of Hidaka Sankei Earthquake

500 gaf

2S0

21. JAN. 1970 AT HIROO HK006

SOOgal.

250 VERTtCAL

SOOgal

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JflL

E-W

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Fig. 4-6

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.

HIROO-EW

UN(M»«>ED NATURAL PERIOD ( SECONDS

)

Response Spectrum Absolute Acc

.

8 HIROO-NS

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i

DAMPING RATIO \

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IV- 3 3
Tab. 4-2 MAIN LIST OF ACCELEROGRAMS

(MAX. ACC. EXCEEDS MORE THAN 100 GAL)

DATE AND NAME

OF EARTHQUAKE

EPICENTER DEPTH

km

STATION

MAX.

ACCELERATION

gal

IBARAGI-OKT 1

' ' 1.1-11 1 i vyXk

E.Q.

Feb. 5, 1964

140.9° E

36.3° N

60.0 KANTO 601 Genken

kishoshitsu

NS 175.0

EW 105.0

NTGATA E 0

May 16, 1964

139.2° E

38.4° N

40.0 NIGATA

701

Kavagishi

cho

apartment

NS 155.0

EV 159.0

E.Q.

May 16, 1968

lA'X AO p

40.7° N

HK 006 Hiroo NS 182.0

EV 165.0

HIGASHIMATSUYAMA

E.Q.

January 1, 1968

139.4° E

36.0° N

50.0 TOKYO

101-2

(TK-024)

E.R.I. NS 106.15

EW 103.68

II II II TOKYO

116-2

(TK002)

I.I.S.E.E. NS 79.45

EW 111.65

HIDAKA-SANKEI

E.Q.

Jan. 21, 1970

143.3° E

42.3° N

60 HK 006 Hiroo NS 412.0

EW 437.0

IV- 3 4
THE UNITED STATES STRONG-MOTION NETWORK:

FIELD OPERATIONS

Richard P. Maley

U. S, Geological Survey

San Francisco, California

ABSTRACT

The national strong-motion instrumentation network operated by the Seismic Engineering

Branch of the U. S, Geological Survey is the system established to record the strong

motions of damaging earthquakes in the United States. From the original 50 C&GS Standard

instriiments installed in the 1930' s, the network has expanded to more than 1300 accelerographs, with 9 different models, located in 35 states and 9 Central and South American

countries.

The network operations section of SEB conducts three interrelated field programs:

instrument installation, routine maintenance, and earthquake record recovery. At the present time the large majority of instruments being installed are self-contained three-component accelerographs that record on 70-mm film. Some remote-sensor accelerographs are

also being located on structures. Routine maintenance intervals have been lengthened from

2 months a few years ago to 4 months, with the exception of a trial area in Los Angeles

where a 6 month interval is now in effect. The higher reliability of modern instrumentation

has made this extended maintenance schedule possible. As the network expands, it may be

necessary to develop a Remote Interrogation System to provide a method of determining the

criterion of the instruments' vital functions by telemetry. After significant earthquakes,

SEB personnel promptly collect and develop earthquake records and attach permanent labels

providing sufficient data for most analyses. The records are then transmitted to the data

management section for further processing.

Key Words: Accelerographs; Earthquake Data; Earthquake Records; Field Stations;

Strong-motion Network.

IV- 3 5
Introduction

In 1932 the United States Coast & Geodetic Survey (C&GS) inaugurated a program of

strong-motion seismological work designed to furnish the engineer and interested others

with data considered essential to the design of earthquake resistant structures (Cloud, 1964X

The responsibility for organizing an instrumentation network to achieve this objective was assigned to the Seismological Field Survey (SFS) in San Francisco, California, the predecessor to the current operating unit, the Seismic Engineering Branch (SEB) of the Office of

Earthquake Studies, U. S. Geological Survey (USGS) . The program was initiated with the

installation of nine low-sensitivity short-period seismographs (accelerographs) in structures selected for special studies by local engineers. Less than 8 months later instruments

installed at Los Angeles, Vernon and Long Beach recorded the disastrous 1933 Long Beach

earthquake. These first useful records of damaging earthquake motions showed amplitudes as large as 0.25g, thus justifying the program and furnishing the impetus for additional efforta

Consequently, the network was rapidly expanded to 50 instruments located principally in the

San Francisco and Los Angeles areas but extending to other seismic regions of the western

United States as well. The principal instrument used in this network was the strong-motion

accelerograph designed by the C&GS in cooperation with the U. S. Bureau of Standards.

During the period 1936 to 1963 the program was marked by a gradual improvement of

instrumentation and methodology and a slow increase in the total number of strong-motion

seismographs. Over this period SFS developed numerous innovations that were subsequently

incorporated into the existing instrumentation. These included among others, the design of

a unifilar accelerometer , a strong-motion displacement meter, and light-tight recording

assembly. Because the original accelerograph was relatively large and required extensive

maintenance at frequent intervals, the network was increased by only 28 units between 1936

and 1963.

In 1963 the first commercially designed accelerograph featuring numerous engineering

and electronic improvements was marketed in California. This instrument overcame many of

the inadequacies of the earlier accelerograph and consequently ushered in an era of substantial expansion in the strong-motion network. With the development of several newer and

less expensive instruments in recent years, the network has continued to enlarge at an accelerating pace. Between 1963 and the present time the number of accelerographs increased

from 70 to approximately 1300. Instrumentation is now located in 35 states, including

Hawaii and Alaska, and in 9 Central and South American countries (Figures 1 and 2) . Because

of the expanded network, field offices have been established at Los Angeles, California, at

Las Vegas, Nevada, and at Columbia, South Carolina.

From its inception, the program has received the cooperation of numerous outside

organizations and individuals. In the early years, housing and facilities were provided by

private and public organizations, and in recent years, a large number of accelerographs purchased by other organizations have been incorporated into the network. In fact, more than

900 of the accelerographs installed since 1963 have been included through the cooperation of

two federal agencies (the Corps of Engineers and the Veterans Administration) , two State of

IV- 36
California agencies (the Division of Mines and Geology and the Department of Water Resources),

and the numerous building departments whose codes require instrumentation at various levels

of high-rise buildings. The building code requirements were initiated in 1965 when the

cities of Los Angeles and Beverly Hills passed ordinances requiring three accelerographs in

newly constructed buildings over six stories high with an aggregate floor area of 60,000

square feet or more, and every building over ten stories high, regardless of floor area.

This requirement became more wide-spread when it was adopted by numerous other communities

in California as a result of being included in the appendix of 1970 Uniform Building Code

(Maley and Dielman, unpublished data) . At this time, approximately one-half of the. total

number of accelerographs in the network are those in high-rise buildings.

Operation of the Network

Instrumentation . The majority of accelerographs in the USGS National Network are the photo

mechanical type. Each instrument contains accelerometers , 'trigger, timer and recorders all

housed in one instrument case. The typical accelerometer is a pendulum with a mirror designed to reflect a beam of light on to translating photographic film or paper. Although

there are six different models of this type in the network, only two are still in general

production, the Kinemetrics SMA-1 and the Teledyne-Geotech RFT-350 (Table 1)

.

1) The Standard C&GS accelerograph records on 152-mm (6 inch) or 304-mm (12 inch)

photographic paper. It is triggered by an oil damped one-second horizontal pendulum that

makes an electrical contact when displaced. The instrument has a fixed operating cycle of

approximately 1 minute and is capable of recording five or six separate events. Several

disadvantages, other than its large size (33 x 50 x 115cm) , include a relatively high

standby current, lack of an internal calibration system, and the need for a darkened room

to change the photographic paper. The instrument has not been manufactured for several

years and is slowly being phased out of the network.

2) The AR-240, the first comparatively modern accelerograph, was designed and marketed

in the United States in 1963. It records on 304-mm photographic paper and is capable of numerous operations because the supply magazine can accomodate rolls of paper 2 3m (75 feet)

to 46m (150 feet) long. The instrument is triggered by a magnetically damped one-second

electrical contact pendulum similar to that in the Standard C&FS accelerograph. The operating cycle is electronically designed to allow continuous operation during an earthquake

and then for an additional 7 seconds after the last pendulum contact. There is an internal

calibration system that may be actij^ated by switches on the outside of the case thus

allowing the easy recording of period and damping at each inspection. Light-proof supply

and take-up magazines provide for the retrieval of records in a lighted room.

3) The MO-2 accelerograph has a fixed 47 second operating cycle and is capable of

registering a maximum of nine distinct events. It records on 35-mm film and is triggered

by an electronic vertical starter that is nominally set to intiate operation at .Olg. The

instrument has no internal calibration system. Records are collected in a light-proof container, although with some difficulty because the film must be mechanically rewound after

recording an event.

IV- 3 7
4) The RFT-250 operates similar to the previously described AR-240 but is somewhat

smaller and records on 70-mm film. Improvements include the elimination of standby current

drain and the incorporation of sealed rechargeable batteries into the instruments base

plate

.

5) The RFT-350 is a re-engineered version of the RFT-250. It possesses basically the

same characteristics except for the addition of an electronic "omnitrigger" that features

a combination of three triggering transducers, two horizontal and one vertical. The

starting threshold level is adjustable but is normally set for .Olg.

6) The SMA-1, a relatively small accelerograph (20 x 20 x 31cm) , records on 70-mm film.

It has the usual features of modern strong-motion instrumentation including light-proof

film supply containers capable of holding 23 m of film, an operate cycle that continues

6 to 20 seconds after the last triggering pulse, internal sealed batteries and a simple

calibration system. The instrument is triggered by a vertical transducer calibrated to

start at .Olg between 1 and 10 Hz. An electrical contact 1-second pendulum starter is an optional feature. Approximately 30 SMA-1 's now in the field are equipped with WWVB radio

receivers that impress real time signals (GMT) on the record. Because the identification

code appears on WWVB once each minute, it is necessary to have a minimum 70-second operating

cycle on these instruments to assure recording of the complete code.

Recently several accelerographs utilizing remotely located unbounded-strain-gage or force-balanced accelerometers have been installed in the network (Table II) . The transducers are encased in small metal boxes ideally suited for mounting in relatively un- accessible locations, such as in the embankment of earthfill dams or on a particular frame

member of a building. The incoming data are transmitted by wire to a centrally located

recorder for amplification and signal filtering. The signals are then sent to a bank of

galvanometers where the deflections of a light beam are registered in the usual manner on translating photographic film. Sensitivity of the components, although adjustable, is

normally s^t to approximately 1.9 cm/g, similar to the 70-mm film recording accelerographs.

The adjustable triggering threshold is set for .Olg and is accomplished by either using a signal from the accelerometers or by separate starters.

As an optional feature, a digital delay memory (DDM) may be incorporated into the

system, thus allowing the recording of data from the transducers prior to actuation of the

recording unit. The DDM continuously monitors the transducer output but records the data

only if the signal proves strong enough to trigger the instrument. Since the data are continuously stored for a short preset interval and discarded if not significant, it is

possible to record the entire earthquake from a time shortly before the P-wave arrival.

The present photo mechanical accelerograph will miss the initial P-wave motion and frequently will not trigger until the S-wave arrives. The DDM functioned well during testing,

and a six-channel unit has recently been incorporated in one of SEB's recorders for field

evaluation.

IV- 3 8
Instrument Installation . Prior to preparation of a site for installation of one or more accelerographs, the SEB

consults with the instrument owner to see that proper facilities are available. Recommended

site criteria include the following: (1) housing that provides ample room for routine

maintenance plus protection from weather, flooding and human interference, (2) a concrete

pad or floor (well anchored to underlying soil or rock if at ground level) , (3) an electrical

outlet for a trickle charger, and (4) the correct shielded interconnecting cable if more

than one accelerograph is to be installed.

After station preparations have been completed, the instrument is set in place by

driving a self drilling anchor into the concrete and bolting the unit securely to the floor.

Both the SMA-1 and RFT-350, the only self-contained accelerographs currently available, are mounted by a single anchor that passes through the center of the base plate. Instruments

installed in a structure are normally aligned with the horizontal accelerometers parallel

and transverse to the structure's axes. Those located along major fault zones as free-field

instruments are situated with the horizontal components parallel and transverse to the trace

of the fault.

A trickle charger is connected to the batteries to assure maintenance of proper charge

level. When no electricity is available or it is not economically feasible to bring in

electrical lines, a solar panel is installed to supply the trickly charging function.

Although modern instruments have an event indicator that reveals whether triggering has

occurred, an external counter is also attached to show the number of operations between

inspections.

When all adjustments have been completed, a test is conducted to record the alignment

of traces and the period and damping of individual accelerometers. This record is returned

to the office for developing, calculation of instrumental constants, and permanent storage

in the station file. After the instrument cover has been replaced, a similar test record

is put on the film so that calibration data will always precede any earthquake record.

The final procedure at installation is to fill out an inspection form providing both

information relevant to the instrument's functioning condition (battery voltage, lamp

voltage, etc) and sufficient details to deteinnine the station location, accelerograph

orientation, local contacts and access to the site (Figure 3) . This information is used by

the technicians to fill out access sheets that are later put in field notebooks so that any

SEB member may independently enter the station to service the instriiment or recover

earthquake records (Figure 4)

.

Photographs are normally taken before leaving the site to show the instrument in

place, its housing, and the major structure associated with it.

Routine Instrument Maintenance

At set intervals one of the SEB technicians visits each accelerograph station and inspects the instrument to see that all components are operating properly and to make any

adjustments and changes required to keep the equipment functioning at its maximum efficiency. The following procedures are carried out during an inspection:

IV- 3 9
1) Check the event counter for possible operations. If it reads anything other

than zero, the potential record is recovered as though it were from an earthquake.

2) If the film supply indicator shows that less than half of the film remains, it

is replaced with a new roll. In any event, the film is replaced at intervals of every two

years or less depending upon the heat and humidity in the instrument room.

3) Manually displace the trigger pendulum to see that it starts the accelerograph.

4) Observe the light traces to see that they are on the collimating lens at the

correct level and that they are spaced at designated locations across the film.

5) Visually check the data traces for damping deflections and free period oscillations.

6) Note whether the time marker solenoids are deflecting properly. Since the advent

of crystal regulated timers, it is seldom necessary to make regular tests of the time rate

accuracy.

7) Measure the charge current with the instrument on standby and the battery load

voltage while the accelerograph is running.

8) Measure lamp current for the nominal operating value.

9) Where several instrioments are interconnected, verify that timing and starting signals are being transmitted between the various units.

10) Check instruments with a WWVB radio receiver for radio reception and an impressed

signal on the time mark solenoids.

After these procedures have been completed, an inspection form (Figure 3) is prepared.

The inspection forms consist of four copy NCR paper so that one duplicate may be left with

the instrument, the second in the station files, the third with the instrument owner, and

the fourth in a master file at SEB's main office in San Francisco.

There are a large number of problems that may be observed during an inspection, but

most of these are minor and inconsequential to the efficient acquisition of strong-motion

data. Some of the more serious malfunctions that cause total instrumental failure or

jeopardize the recording system are summarized in the following paragraphs.

1) Loose instrument mounting. Frequently the anchors imbedded in concrete loosen

after installation resulting in a loose instrument within the next several months. Retightening the anchor nut eliminates this problem.

2) Electronic breakdown. A number of solid-state failures may occur including those

within circuit boards regulating the trigger and operate cycle. In the first instance, the

instrument will not start and in the second instance, the instrument will not turn off

after having been triggered. Other failures have occurred in circuit boards governing the

motor drive, time pulse generator, and calibration systems. The inspecting technician

carries a supply of these components in the field and can readily replace the malfunctioning

units when they are discovered.

3) Power loss. Power loss has been a major long-standing problem with strong-motion

accelerographs but of a diminishing nature since small trickle chargers were installed

on the batteries after SEB's experiences in the 1971 San Fernando earthquake (Maley, 1971).

Some of the problems that remain are the occasional breakdown of chargers, normal battery

IV-40
failure, loss of electrical power and subsequent charge current, and the corrosion of

battery terminal connectors.

4) Mechanical maladjustments. The 1-second horizontal pendulum starters may become

offset and if this is sufficient to make the electrical contact, the instrument will trigger and not turn off until the batteries have gone dead. The electronic vertical starters

somethimes sag until they come to rest on their lower stops.

Other problems in this category include trace drift that may result in total loss of

one or more channels of data, the loss of time mark solenoid deflections resulting in the

omission of time signals on the record, and complete or partial failure of the film drive

mechanism.

5) Environmental problems. Excessive heat and humidity may cause impairment of the

normal instrumental operation. Occasional flooding has occurred, resulting in total disablement as well as an expensive repair bill. Human interference is also a problem,

sometimes merely from the curious but at other times from vandals bent on destruction. In

one instance, a locked accelerograph building was broken into and the instrument pried off

its mounting and then thrown into an irrigation canal.

Most of these problems can be handled immediately during the inspection, but when

field repairs are not expedient, the instrimient is removed and taken to a SEB shop for

renovation.

Until the mid 1960 's the national network consisted chiefly of Coast and Geodetic

Survey Standard accelerographs and a lesser number of AR-240's. At that time it was particularly important to see that each instriament was kept in good operating condition

because there were relatively few stations to record any one earthquake. Since the existing accelerographs were slightly less reliable than those developed in recent years, a maximum inspection interval of 2 months was considered necessary to obtain a high data return. As the network was rapidly enlarged by the influx of newer and more reliable instruments, the inspection period was lengthened to 3 months. Even with this quarterly

scheduling, it was impossible for the limited SEB staff to keep up with the accelerating

number of new installations and inspections required.

After a 1974 study of maintenance procedures in the network, that was by now heavily

weighted with highly reliable instriomentation, it was determined that the service interval

could be further lengthened to 4 months with no significant decrease in operating efficiency.

The entire network is now being maintained at 4-month intervals except where chargers have

not yet been installed and for more than 100 buildings in the Los Angeles area a test

is being conducted using 6-month inspections. This excludes some more critical or troubleprone stations that are serviced more frequently. Should this test prove successful, SEB

intends ultimately to convert the entire network to a 6-month maintenance schedule. Because a large part of technician time is expended in simply traveling to the stations, the

less frequent inspection interval provides for more time to be used in servicing the

instrument, thus resulting in better performance. This trend was indicated by the 1974

study of maintenance procedures.

IV- 41
SEB is striving to achieve a 95% success rate in keeping the instrumentation operational. This seems to be a desirable goal when one considers the unpredictable recurrence

interval for strong earthquake motion at any particular location. An instrument may re- main virtually idle and not record important data for 20 years, but when an earthquake does

occur, if the instrument is not operational, a unique opportunity for recording damaging

motion at that site may be lost. The extremes in significant recording intervals are

shown by Long Beach, where an important record was obtianed 8 months after the instrument

was installed (only several records of minor importance have been obtained subsequently)

,

and San Diego, where no significant record has been obtained in the 41 year history of the

station. The necessity of keeping instriiments in good condition in the eastern United

States is of even greater importance because the probability of a damaging earthquake

occurring in any period of time is considerably less than in California.

A longer inspection interval will help keep pace with the maintenance of a rapidly

expanding network for some time, but if in the next several years the number of accelerographs doubles or triples, some alternative methods must be considered. One that perhaps

offers the most promise for the future is the development of a Remote Interrogation

System (RIS) to determine instrument operating capability by a telemetry link to a central

control unit.

The development of a RIS could allow maintenance personnel to remotely interrogate

any accelerograph for the condition of vital instrument functions such as proper triggering,

battery voltage under load, amount of film remaining, and the number of events recorded.

The interrogation could be preprogrammed to cover any set of stations or manually operated

to select an arbitrary group of stations. SEB technicians may effectively inspect the

network at frequent intervals and yet make actual on-site field maintenance visits only

when serious malfunctions are observed. It is anticipated that other than such required

maintenance trips, each instrument would be thoroughly serviced annually.

A second function of the RIS would be to querey the event counter of any selected

group of instriaments after a local earthquake. Although most felt earthquakes are not of

engineering significance, records at a particular site may be of special value to some

enigneers, geologists, or seismologists. The event-counter interrogation would simplify

the recovery of such records and would, to some extent, define the time of occurrence of

minor earthquake records that are collected during routine servicing.

Besides its obvious advantages, the RIS would pay for itself over a period of years

by reducing the mean annual maintenance costs, primarily that of travel and manpower. As

the system become operational, it would allow an increase in the network size while permitting more effective use of personnel.

Earthquake Record Recovery

Post earthquake procedures in the handling of records cover three broad phases:

1) the actual retrieval of records from individual instrument sites, 2) the photographic

development of records, and 3) the labeling of the records with sufficient data for most

routine analyses. Because of the large number of small earthquakes that occur in California,

this is nearly a continual process although at a substantially lower level than after

IV- 4 2
a significant event.

A casual review of the number of records obtained during some recent earthquakes shows

the magnitude of the recovery and documentation that will be required during future large

shocks. In 1968, 114 records were obtained from a magnitude 6.5 earthquake that occurred

210 km from Los Angeles. During the 1971 San Fernanco earthquake, 241 accelerograms and

at least 100 aftershock records were recorded. The magnitude 5.9 Point Mugu earthquake in

1973 produced 318 accelerograph records. Even the recovery of lesser interest records,

such as from the Point Mugu earthquake, presents a formidable task.

The interrogation system proposed above will be of considerable assistance in record

recovery as it may be used to monitor instrument operations after locally felt earthquakes

that otherwise are not damaging and of relatively little significance. For instance, frequent smaller earthquakes are felt in the Los Angeles area, but it is impossible to determine which instruments have been triggered without actually visiting each site.

Considering the present size of the network in southern California, an earthquake in

the magnitude 6.5 range near Los Angeles would result in 700 original records and at least

200 aftershock records. It is estimated that some 3,500 feet of data would be simultaneoulsy obtained on a variety of recording media, i.e., 152-mm and 304-mm photographic

paper and 35-mm and 70-mm film. To collect, develop, and label such a large set of records

from a single event is a project of staggering proportions, particularly when a large

amount of time must be allotted to dealing with the inquiries from engineers, scientists,

and public officials. For example, it took nearly 4 weeks effort by the entire SFS

staff just to collect and develop all the records obtained during the Sanfernando earthquake .After an earthquake, SEB field personnel proceed to the spicentral area as rapidly

as possible to retrieve records and return the instruments to their pre-earthquake condition. They carry sufficient provisions needed for extensive recovery, especially loaded

film and paper magazines to replace those that have substantially reduced supplies because

of the earthquake operations. Upon entering the station the technician will check the

signal counter, put calibration data on the record, and advance sufficient film into the

take-up magazine to assure protection from light when the instrument is opened. The record is then removed and labeled (etched on film or written on paper) with the station name,

accelerograph serial number, date of record recovery, and date of earthquake, if known.

Finally, the accelerograph is restored to its normal operating condition and the counter

reset to zero. The records are returned to the local SEB office for photographic processing. In the

past, this has been accomplished by sight developing because of the instruments' variable

lamp intensity and the possibility of accidental fogging during record recovery. Considering the potentially large number of records from future earthquakes, some effort is

being expended to provide automatic developing procedures. A paper accelerogram processor

is being tested that will develop, fix, and dry the records in one operation. A 70-mm film

dryer is now in operation, and the use of complete film processing systems are being

investigated.

IV- 4 3
After development, the records are temporarily labeled by affixing a giammed sticker

with the station name, instrument number, and date of the earthquake. Pemanent labels are

later produced by typing the following information on matte-finish mylar plastic: station

name, permanent station number, instrument type and serial number, data of earthquake in

both GMT and local time, and the physical constants of the individual components including

orientation, sensitivity, period and damping (Figure 5) . These labels are spliced directly

to the original record, normally just in front of the pre-earthquake calibration data. The

records are then transferred to SEB's data management section for the production of distribution copies and subsequent processing and analysis.

ACKNOWLEDGEMENTS

The strong-motion program conducted by the Seismic Engineering Branch, U. S. Geological

Survey is supported by National Science Foundation grant (NSF-CA114) . The author is indebted to the staff of the Seismic Engineering Branch who provided numerous valuable suggestions during the preparation of this paper.

IV-44
References

1. Cloud, W. K. , 1964, The cooperative program of earthquake investigations, in Carder,

D.S., ed.. Earthquake investigations in the western United States, 1931-1964: Dept. of

Commerce, Coast and Geodetic Survey Pub. 41-2, p. 3-4.

2. Halverson, H. T. , 1969, Some recent developments in strong-motion seismographs:

Teledyne-Geotech Pub., Monrovia/ California, 32 p.

3. Kinemetrics Inc., 1974a, SMA-1 strong-motion accelerograph: specification sheet,

Kinemetrics, San Gabriel, California 2p.

4. Kinemetrics Inc., 1974b, Preliminary data sheet, CR-1 central recorder, Kinemetrics,

San Gabriel, California, 3p.

5. Maley, R. P., 1971, A statistical summary of accelerograph performance during the earthquake of February 9, 1971, iri Hudson, D. E., ed. , Strong-motion instrumental data on the

San Fernando earthquake of February 9, 1971: California Institute of Technology and

Seismological Field Survey, NOAA, Dept. of Commerce, p. 54-76.

6. Nimbus Instruments, 1974, TMA-100 dynamic recording system, West Sacramento,

California, Nimbus Instruments, 4p.

7. Teledyne-Geotech, 1973, System description RRA-100, Garland, Texas, Teledyne-Geotech, 7p.

IV-45
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IV- 50\
U. S. GEOLOGICAL SURVEY

SEISMIC ENGINEERING

DATE

STATION 11611 San Vicente Blvd.

ADDRESS Brentwood

CONTACT Fad Wiley

POSITION Building Engineer

STATION # 875

INSTRUMENTS SM-1 s/ n 1414 Grnd

s/n 1415 6th

CITY Los Angeles

PHONE 826-0889

KEYS 1

L- S 05 W

V- Down

s/n 1416 Roof(ll) T- S 85 E

TOPO SHEET Beverly Hills C0-0RD34.O5 N 118.25 W

THOMAS MAP P.# LA-41

OWNER CODE LA

CROSS STREETS S. Vicente - Bringham

ORIENTATION VERIFIED Yes

DESCRIPTION Enter pkng. from rear. Ground - enter acc. room from

driveway to ground level pkng. (San Vicente side). 6th - en central

core east of el. Roof - el. to 10, NW stairs to roof, 3rd door

in penthouse.

1-405

Son Vicente

r

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fa]

occelerogroph

So n Vicente

Figure 4. Typical accelerograph station access sheet.

IV- 51
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IV- 5 2
Strong-Motion Data Management

by

Christopher Rojahn

U. S. Geological Survey

San Francisco, California

ABSTRACT

The Seismic Engineering Branch (SEE) , of the Office of Earthquake Studies, U. S.

Geological Survey, is funded by the National Science Foundation and is responsible for the

development and maintenance of a national network of strong-motion instrxaments and for the

processing, management, and dissemination of data obtained from those instruments. Data

management is central to the entire strong-motion program; it serves as a focal point for

the functions of archiving the records, processing the data, and disseminating both the

data and information about the program to the user community. In the archival phase, all

records are stored by station and cataloged both by event and by station. In data processing, all significant ground and basement level records are digitized after which the

raw digitized data is used to generate the following: uncorrected acceleration timehistories; velocity and displacement time-histories; and various forms of frequency domain

spectra. Both SEB and the Environmental Data Service of the National Oceanic and

Atmospheric Administration are involved in the data and information dissemination operation. Each organization distributes data, whereas SEB is solely responsible for the

dissemination of information about the strong-motion program. Various U. S. Geological

Survey professional papers and circulars are the primary media through which the latter

function is accomplished.

Key Words: Accelerographs; Data Processing; Earthquake Records; Strong-motion Data

IV- 5 3
Introduction

The Seismic Engineering Branch (SEB) of the Office of Earthquake Studies, U. S. Geological Survey (Department of Interior) , is responsible for the development and maintenance

of a national network of strong-motion instruments and for the processing, storage, and

dissemination of data obtained from those instruments . In order to carry out these

responsibilities, SEB concentrates its efforts on the following activities: program

management, network design, network operations, data management, and research and applications. Data management, the subject of this paper, is central to SEB's entire strongmotion program. It serves as a focal point for the functions of (1) archiving the records,

(2) processing the data, and (3) disseminating both the data and information about the program to the engineering seismology research community, the structural design community,

and regulatbry agencies at the Federal, State, and local levels. All three functions are discussed in detail in subsequent portions of this paper.

Background Information

SEB's data management functions were first outlined in a proposal submitted to the

National Science Foundation (NSF) in July 1974. During the year prior to that time, SEB

operated under general NSF funding in a transitional stage during which its responsibilities and functions were formulated and defined in detail. Before that, (i.e., up

until May 27, 1973, when SEB was formed as part of the U. S. Geological Survey), the

office existed as the Seismological Field Survey (SFS) in the National Oceanic and Atmospheric Administration (NOAA) of the U. S. Department of Commerce. In general, the SFS had

fewer responsibilities, a smaller staff, and substantially less funds. In particular, the

data management function was the responsibility of NOAA's Environmental Data Service (EDS).

As a result of the July 1974 proposal, NSF awarded SEB a grant of $700,000 annually

to develop and assume primary responsibility for the U. S. strong-motion program. The

funding was approved in principle for five years and will be increased annually in

accordance with a nominal inflation factor. Of the total annual amount, $240,000 has been

allocated for data management (FY 1975) . This level of funding has enabled SEB to acquire

three additional staff members in data management. Consequently, in addition to the project chief, there are now six persons working on data management—one geophysicist, one mathematician, one computer specialist, two physical science technicians, and one clerk

typist. In contrast, there are 16 persons assigned to other SEB projects (22 staff members

in total)

.

In its present form the national network of strong-motion instruments contains

approximately 1300 accelerographs located primarily in California but also throughout the

remainder of the seismically active areas of the U. S. The instruments are owned by

Federal, State, and local agencies, universities, private firms and individuals, and other

independent groups. SEB acts as the primary coordinator for the development and maintenance of the entire network and assumes overall responsibility for record collection, record

archiving, data analysis and data dissemination.

At present, the vast majority of instruments in the network are triaxial optically- 54
mechanical accelerographs that record three strong-motion traces per record. Among the new

installations, however, are remote recording systems that record up to 12 strong-motion

traces per record. In addition, direct digital recording systems are also expected to be

installed in the near future. These differences in the types of recording systems and the

number of traces per record have made it necessary to design a data management system that

can handle a variety of data forms.

Archival System

As of December 31, 1974 approximately 3200 strong-motion records had been obtained

from the national network of strong-motion instruments. Although the first record was generated in 1932, the vast majority have been obtained since the mid-1960's, when the

number of instruments in the network began to increase rapidly. Many of the records are

on 6~inch and 12-inch (152-mm and 305-mm) paper, a few are on 35-mm film, but most are on

70-mm film. It is expected that most of the records obtained in the near future will be

on 70-mm film, a substantial number of others will be on various sizes of paper and film

(12-inch, 35-mm, and others) , and a few will indoubtedly be on magnetic tape.

The records are archived by station, which is defined as the geographic location

at which an instrument (s) is (are) located (e.g. a building housing three instrximents constitutes one station) . There are one or more storage containers (depending on the number

of existing records) reserved for each station. The 35-mm and 70-mm film records are

stored in 2 7/8-inch (73-mm) -high, 3 3/4-inch (95-mm) inner diameter light-weight metal

canisters; the larger film and paper records are stored in 15-inch (381-mm)-long, 2 7/8-

inch (73-mm) inner diameter medium-wall cardboard tubes. Each container is labeled with

the station name (address or structure title) , assigned station number, and date of

event (s) of records stored therein. Measures have been taken to ensure that the archival

room is as fire-proof as possible.

Prior to permanent storage, the records are labeled, and high-quality digitizable

mylar copies are made of all those considered significant (in general, ground or basement

level records having a maximum acceleration greater than, or, equal to O.lOg).

Beginning in middle or late 1975, one mylar copy will be sent to each of the following:

the SEB branch office in Los Angeles; the California Institute of Technology in Pasadena;

the Environmental Data Service, NOAA, in Boulder, Colorado; and the instrument owner. All

other records are archived without being copied unless there is a specific request to do

so.

One other important part of the archival process is the preparation of an "event and

station information card" (figure 1) for each strong-motion record. The card is prepared

as soon as a record is received and contains all pertinent station and event information,

instruments constants, maximum accelerations and epi-central distance. One copy is

filed in a station file, and a duplicate is placed in an event file. Eventually, the

information from both files will be published in catalog form and will be made available

on a computerized data file that can be queried by remote terminal.

IV- 5 5
Data Processing System

SEB's data processing system is evolving from in-house developments , but much of it is

patterned after the methodology developed by the California Institute of Technology (CIT)

during the digitization and analysis of the February 9, 1971 San Fernando earthquake

strong-motion records (1). Prior to the award of the 1974 NSF grant, SEB's data processing

system was smaller in scope though similar in form. Presently, it is being expanded to

handle a large volume of records in a relatively short time.

As a matter of course, SEB plans routinely to process all significant ground and basement level records. The processing of each record can be subdivided into several phases:

record digitization; conversion of raw digitized data to uncorrected data; conversion of

uncorrected data to corrected data and the generation of velocity and displacement timehistories; and the generation of various forms of frequency domain spectra. Phases two

through four are accomplished through the use of three computers, a CDC 6400, CDC 6600, and

a CDC 7600, all of which are located in Berkeley, California. All computer programs are

submitted in batch or time-share mode either at Berkeley or via one of the remote teminal

systems at the U. S. Geological Survey (USGS) computer center in Menlo Park, California.

The first phase of data processing (i.e. record digitization) is not carried out inhouse. Instead, digitization services are being procured from the following external

organizations: Dynamic Graphics of Berkeley, California; I/O Metrics of Sunnyvale, California; and EDS/NOAA of Boulder, Colorado. Dynamic Graphics uses a manual film-plane digitizer and is presently able to digitize both enlarged film and normal-sized paper records

on a routine basis. The system is particularly appropriate for those records that cannot

be digitized automatically, that is, those records that are not well defined, are noncontinuous, and have overlapping traces. Normally, the firm is requested to digitize all

peaks, valleys, and inflection points at a minim\am rate of 50 points per second. I/O

Metrics has developed an automatic, laser-beam, trace- following, film-plane digitizer

and has satisfactorily demonstrated its capabilities in a limited trial. Film records

may be digitized directly, whereas paper records must be reproduced on 70-mm film before

digitization. The system is especially appropriate for well defined, continuous, non- overlapping traces, and is notable for its high processing speed potential (the firm

estimates that it can accurately digitize an average-sized strong-motion record in half

an hour). Normally, records are digitized in increments ranging in width from 25 to

50 micros. EDS uses an automatic faster-scanning, equal- increment digitizing system

(VISICON) which is capable of scanning a 12 x 25-inch (305 x 635-mm) record at a maximum

rate of 200 samples per inch (approximately 79 samples per centimeter) . Records digitized on the system are currently being evaluated by SEB. It is anticipated, however,

the system will be satisfactory for processing both paper and enlarged film records. In

total, the three external organizations collectively provide SEB with the capability for

digitizing a large volume of records in a relatively short time.

SEB requires that each external organization si±imit its digitized data on magnetic

tape or punched cards in a preselected format, and that it be accompanied by a full-scale

plot of the data (a plot at 3x's scale is also frequently required for film records). Each

IV-5e
plot then xindergoes a quality control check whereby it is visually compared to the original

record (or 3x's scale copy) with the aid of a low-power magnifying lens. If the plot of

the digitized data appears to be truly representative of its analog counterpart the

digitized data are accepted for subsequent phases of processing. If not, the original record is redigitized.

In the second phase of data processing, the raw digitized data are converted to what

is referred to as "uncorrected data" (2). At this time, all paper or film distortions are

removed, and acceleration and time are cross-correlated in order to express acceleration as

a function of time. The final product is permanently retained on either magnetic tape or punched cards (in the near future, such data will also be stored on microfilm ) , on a digital listing, and as a time-history plot.

During the third phase, uncorrected data are converted to "corrected data" (3) , and

velocity and displacement curves are generated. Instrumental and baseline corrections are

first applied to obtain true ground acceleration (corrected data) . Single and double

integration processes are then used to generate velocity and displacement time-histories.

The data are permanently retained on either magnetic tape or punched cards (in the near

future, such data will also be stored on microfilm ) , on digital listings, and in plot

form.

During the fourth and final phase of data processing, various forms of frequency

domain spectra are generated. Along these are the spectra routinely calculated and

plotted by CIT (4, 5) during the processing of the San Fernando earthquake strong-motion

records: maximum relative velocity response spectra (figure 2); relative displacement,

pseudo velocity response, and pseudo acceleration spectra plotted on tripartite paper

(figure 3) ; and Fourier amplitude spectra (figure 4) . In addition, SEB has developed two

other forms of plots (6) that give substantially more insight into the nature of strongmotion: relative velocity response envelope spectra (figure 5) , and time-duration spectra

of the response envelope (figure 6) . All five forms of spectra are permanently retained

on digital listings and in plot form.

Data And Information Dissemination System

Information about the SEB strong-motion program is disseminated solely by SEB, whereas

the actual strong-motion data are disseminated both by SEB and EDS. EDS is the distributing agency for the uncorrected and corrected data generated by SEB. Such data are made

available in punched card form and on magnetic tape (7 or 9-track) . EDS also provides

analog copies of strong-motion records on 70-mm film clips (approximately 8x's reduction)

and on 35-mm film reels (12 's reduction) and is being requested to develop the capability

to produce blue-line full-scale analog copies. Beginning in middle or late 1975, SEB

will publish routine data reduction and analysis results in a series of U. S. Geological

Survey professional papers. It is expected that each professional paper will cover one calendar year and will contain the following data for all significant ground and basement

level records generated in that year: uncorrected and corrected acceleration timehistories; velocity and displacement time-histories; and the five forms of spectra desIV- 5 7
cribed in the section on data processing. In addition, SEB also distributes, upon request,

blue-line analog copies of all records not made available through EDS.

Information about the strong-motion program is disseminated primarily through the

following publications: the "Seismic Engineering Program Report," a USGS circular published quarterly; "Strong-Motion Accelerograph Station Locations Listing," a USGS circular

published annually; "Strong-Motion Accelerograph Station Descriptions" (tentative title)

,

a series of USGS professional papers; and "Catalog of Strong-Motion Records," tentatively

planned as a joint USGS/NOAA publication (to be issued as a USGS professional paper) . All

circulars are issued free of charge on request (the current mailing list contains names and

addresses of approximately 2300 recipients) . The professional papers receive a limited

free distribution but can be purchased for a nominal fee.

The first issue of the "Seismic Engineering Program Report" (7) was published in

December 1974 as U. S. Geological Survey Circular 713. It contains a listing of 1972 and

197 3 accelerograph records and a description of the then-current status of the SEB strongmotion program. The second issue, scheduled for publication in the spring of 1974, contains a listing of 1974 accelerograph records, two articles by SEB staff members (one on

the Latin American and Caribbean strong-motion programs and one on recent developments

in strong-motion instrumentation) ,and notes on record corrections and the availability of

strong-motion data. Future issues will contain similar information with all listed data as

current as practicable.

The first issue of the "Strong-Motion Accelerograph Station Locations Listing" is

scheduled for publication in mid-1975. It will contain the following data for all stations

installed in the national network prior to January 1, 1975: station nxamber, location,

country, structure type/size, instrument location (s) , coordinates, and source(s) for data.

The list will be routinely updated in the "Seismic Engineering Program Report." The

second issue of the listing will contain the same information for all stations installed

prior to January 1, 1976 and will be published in early 1977. Similar listings will be

published annually thereafter.

The content and format for "Strong-Motion Accelerograph Station Descriptions" have not

yet been firmly established. It is probable, however, that the descriptions will contain

the following: location data including the geographic coordinates, road and topographic

maps, and an exterior photograph; geologic data including a brief seismic history of the

area, the proximity of faults, if any, a verbal description of the regional and local geology, cross-section (s) , formation descriptions, and depth to water table; and soils data,

when available, including boring logs and measurements of density, standard penetration

values, and P- and S-wave velocities . If there is a structure at the site, each station

description will (tentatively) include: typical plan view(s) indicating lateral force

resisting elements, orientation, dimensions, and the center of mass and center of rigidity

of each floor or mass element; typical section (s) of the full structure showing dimensions; general design information (e.g., mode shapes and frequencies); and the source where additional information may be obtained (e.g., the design engineer).

It is planned that the "Catalog of Strong-Motion Records" contain a listing of all

IV- 5 8
accelerograph records obtained from the national network of strong-motion instruments since

its inception in 1932. Records will be listed both by event and by station showing epicentral distance, maximum acceleration whenever available, and various data sources. The

frequency of publication has not yet been established, but is expected to be annually,

biannually, or even less frequently.

Summary

The Seismic Engineering Branch (Office of Earthquake Studies, U. S. Geological

Survey) is funded by the National Science Foundation and is responsible for the development

and maintenance of a national network of strong-motion instruments and for the processing,

management, and dissemination of data obtained from those instruments. Data management is

central to the entire strong-motion program; it serves as a focal point for the functions

of archiving the records, processing the data, and disseminating both the data and information about the program to the user community. In the archival phase, all records are

stored by station and cataloged both by event and by station. In data processing, all

significant ground and basement level records are digitized after which the raw digitized

data is used to generate the following: uncorrected and corrected acceleration timehistories; velocity and displacement time-histories; and various forms of frequency domain

spectra. Both SEB and EDS are involved in the data and information dissemination operation.

Each organization distributes data, whereas SEB is solely responsible for the dissemination

of information about the strong-motion program. Various U. S. Geological Survey professional papers and circulars are the primary media through which the latter function is

accomplished.

Acknowledgments

The author gratefully acknowledges Chuck Knudson, Dick Maley, and Fritz Matthieson

for their help in reviewing and editing this manuscript.

IV- 59
References

1. Trifunac, M.D., and Lee, V., 1973, Routine computer processing of strong-motion

accelerograms: California Institute of Technology report N. EERL 73-03, 360 p.

2. Earthquake Engineering Research Laboratory (EERL) , 1969, Strong-motion earthquake

records, Volume I-uncorrected accelerograms: California Institute of Technology

report No. EERL 70-20, part A, 203 p.

3. Earthquake Engineering Research Laboratory (EERL) , 1971, Strong-motion earthquake

records, Volume Il-corrected accelerograms and integrated ground velocity and

displacement curves: California Institute of Technology, report N. EERL 71-90,

part A, 321 p.

4. Earthquake Engineering Research Laboratory (EERL) , 1972, Analyses of strong-motion

earthquake accelerograms. Volume Ill-response spectra: California Institute of

Technology report No. EERL 72-80, part A, 272 p.

5. Earthquake Engineering Research Laboratory (EERL), 1972, Analyses of strong-motion

earthquake accelerograms. Volume IV-Fourier amplitude spectra: California

Institute of Technology report No. EERL 72-100, part A, 164 p.

6. Perez, Virgilia, 1974, Time dependent spectral analysis of thirty-one strong-motion

earthquake records: U. S. Geological Survey open-file report, 141 p.

7. U. S. Geological Survey (USGS) , 1974, Seismic engineering program report,

October-December, 1974: U. S. Geological Survey circular 713, 19 p.

IV-60
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IV-65
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IV-66
BRIEF REVIEW ON LIQUEFACTION

DURING EARTHQUAKES IN JAPAN

by

Eiichi Kuribayashi

Chief of Earthquake Engineering Research Section

Public Works Research Institute

and

Fumio Tatsuoka

Research Engineer

Earthquake Engineering Research Section

Public Works Research Institute

ABSTRACT

To correlate the actual liquefaction phenomoia and site conditions , a literature survey

about the liquefaction phenomena caused by earthquakes during the last century in Japan

was performed. A liquefaction distribution map of Japan and the regional maps of Kanto,

Nobi and Hokuriku are presented and the factors related to liquefaction are discussed.

During the last century liquefaction in sub-soils have been observed at some hundred sites

during 44 earthquakes where the sites were limited to alluvial deposits and reclaimed lands.

Furthermore, it was found that liquefaction occurred repeatedly in different earthquake

zones. The estimated JMA intensity scale factor, at the liquefied sites was more than five

which means a maximum acceleration of 80 to 250 gals. The extent of the liquefied zones are limited , depending on the magnitude of the earthquake.

Key Words: Alluvial Deposits; Earthquakes; Epicentral Distance; Liquefaction; settlement;

Soils.

y-1
Introduction

Most structural damage, caused by settlement or inclination due to liquefaction of

saturated sandy subsoil, has inevitably occurred during major earthquakes. In accordance

with the prediction of liquefaction potential of subsoils, a number of research efforts

have been expended by means of laboratory tests and in-situ tests. Meanwhile, correlations

between the actual liquefaction phenomena and the site conditions has rarely been investigated. Recently, case reviews on the correlation of liquefaction have been performed at

several sites in Japan (Miyabe, 1933, Kishida, 1969, Ishihara, 1974 and lida, 1974). Because many interesting facts have been obtained from these reviews, more reviews on this

subject are expected. This report contains the results of a literature survey about

liquefaction phenomena caused by earthquakes during the last century in Japan.

Liquefaction distribution maps since 1972 are presented first and the factors related

to liquefaction are then discussed. In the literature on earthquakes, which caused damage,

there contains much description of the liquefaction of the sub-soil. However, most of the

descriptions before 1868, the identification of liquefaction and its characteristics,

have been scientifically qualified. In the description, evidences of liquefaction in the

sub-soil are given from i) water spouting with sand or mud from wells or cracks in the

ground, ii) sand boils or sand volcanoes, iii) excessive settlements of heavy structures

placed on sand layers, and iv) uplifting of wooden piles from rice paddy, or of caissons

under construction from river beds. About 150 papers or reports have provided information

on liquefaction during 44 earthquakes, as listed in Table-1.

Liquefaction Distribution Maps

Forty-four earthquakes have caused liquefaction of the sub-soil in Japan during the

past century. The epicenters of the earthquakes and locations where liquefaction has

occurred is shown in Fig. 1. The characteristics of the earthquakes, which include date,

magnitude, focal depth, epicenter location and the number of the deaths are also listed in

Table 1. Large earthquake magnitudes cause liquefactions in wide areas except for several

cases, as shown in Fig. 1. Among the large scale earthquakes were No. 4 (Nobi) , No. 12

(Gono) , No. 15 (Kanto) , No. 22 (Nishi-Saitama) , No. 34 (Fukui) and NO. 42 (Niigata) as

shown in Table 1 which caused wide distribution and violent liquefaction of the sub-soil.

Moderate scale earthquakes. No. 6 (Shonai) , No. 8 (Rikuu) , No. 16 (Tajima) , No. 31 (Tonankai) , and No. 44 (Tokachi-oki) caused local violent spouting of water or a number of large

sand volcanoes

.

Microscopic features of the regions of liquefaction are closely distributed as shown

in Figs. 2, 3 and 4 which depict Kanto, Nobi and Hokuriku regions, respectively.

Kanto Region

In the Kanto region during the earthquake No. 5 (Tokyo) , spouting of water with dark

blue or black fine sand from cracks of subground were observed in many places. In Fig. 2

the sites of liquefaction are denoted by solid circles with numeral characters of 1 to 18,

which are distributed only in alluvial areas and close to the epicenter. About 30 years

after this earthquake, the Kanto region was subjected to earthquake No. 15 (Kanto) , which

V-2
was one of the most disasterous earthquakes in Japanese history. Sites of liquefaction,

during this earthquake, as denoted by hollow triangles in Fig. 2 are very widely scattered

in the alluvial plains in the Kanto region. It is interesting to note that many liquefied

sites are concentrated on the original winding river course of the existing Tone river located in the north east of Tokyo. About three hundred and fifty years ago, the estuary

of the Tone River was linked by man to the Pacific Ocean. The original Tone River had

flowed into Tokyo Bay, depositing a very deep and soft sand layers. Because the river

course was modified, a higher potential of liquefaction was retained in the original site

in comparison with the ordinary site in Kanto plain. In addition to the earthquakes described above, earthquakes No. 7 and No. 22 (Nishi Saitama) during this century has

caused liquefaction of sub-soil in this area. In addition to the Kanto area, sand boils,

sand volcanoes, uplifting of wooden piles and caissons during the earthquake No. 15

(Kanto) were reported as shown in Fig. 2, where representative liquefied sites are denoted

by characters a, b, ... k. For example, the site a is the reclaimed land in Kawasaki City,

where water with dark blue and grey fine sane ejected from many cracks. At site b near the

epicenter, seven wooden piles uplifted about 60 cm from the rice paddy during the quake.

These piles were used as bri-dge piles about 800 years ago over the original Sagami River.

At site c many caissons for a bridge foundations under construction uplifted from the

river bed of the Sagami River. More details of liquefaction situations are reported in

another paper (Kuribayashi , Tatsuoka and Yoshida, 1974)

.

Earthquake No. 22 (Nishi Saitama) in 1931 also induced liquefaction at many locations

along the original and existing courses of the Tone River. The liquefaction sites during

this earthquake, are denoted by hollow circles as shown in Fig. 2. Severe water spouting

conditions were observed at sites around Fukiage City, located about 20 km east from the

epicenter. Large amounts of water, with dark blue and grey fine sand, were ejected about

30 cm high from several hundred of these sand volcanoes. Furthermore, the ground surface

of these liquefied sites settled about 15 cm and the spouted water covered the ground to a

depth of about 15 cm.

Nobi (Mino-owari) Region

The Nobi Region has been inflicted by four violent earthqukaes during the past century,

which induced liquefaction at many sites as shown in Fig. 3. The most violent was No. 4

(at Nobi) in 1891 which caused wide spread and violent liquefaction of the Nobi Plain and

Fukui Plain. These locations are shown by solid triangles in Fig. 1 and by hollow triangles in Fig. 3, respectively. The liquefied sites were limited to the alluvial soft

deposits along the Kiso, Nagara and Ibi Rivers in Nobi Plain and to the sites along the

river in the Fukui Plain which is to the north of the epicenter. It should also be mentioned that the earthquake No. 34 in 1948 (Fukui) also induced liquefaction in the entire

Fukui Plain as shown in Fig. 1. One of the most violent water spouting conditions occurred

during earthquake No. 4 and is denoted by a character "a" in Fig. 3. During this earthquake near the Shonai River, water with sand ejected over 2 m high from wells and deposited

sand on the roofs of nearby houses. Also at the site "b", on the right bank of the Shonai

River to the south of the site "a", twelve hundred sand boils were observed. The extent

V-3
of liquefaction can be evaluated in that 40 km^ of rice paddies and fields in the

Prefectures of Aichi and Gifu, were impossible to cultivate due to sand boils, necessitating repairs and reconstruction.

Eighteen years after this earthquake, earthquake No. 12 (Gono) caused liquefaction at

many locations in the northwestern part of the Nobi Plain where liquefaction was previously

caused by earthquake No. 4. Many areas liquefied during this earthquake, and are denoted

by hollow circles as shown in Fig. 3. These areas are tightly distributed and are sur- rounded by the Ibi , Nagara and Kiso Rivers

,

Earthquake No. 31 (Tonankai) , during the end of World War II caused liquefaction in

the coastal region which has soft alluvial deposits. These locations are between the

Cape Omaezaki and the Lake Hamana and in the south coastal part of Nagoya City, which had

been reclaimed during the past hundred and fifty years before this earthquake. The

liquefied sites described above are expressed by hollow triangles as shown in Fig. 1 and

solid triangles as shown in Fig. 3. In the south coastal region of Nagoya City damage to

wooden houses such as settlement and inclination was caused by a very large quantity of

ejected sand and water. AlsOfmany sand boils and water spoutings were observed at the

sites of severely damaged aeroplane factories at Nishi-inaei town, denoted as character

"c" in Fig. 3. During earthquake No. 32 (Mikawa) , violent boiling of sand and water was

again observed at the sites of these same factories.

In genera], the characteristics of earthquakes of large magnitude, but with offshore

epicenters such as the earthquakes No. 31 (Tonankai) , No. 33 (Nankai) and No. 44 (Tokachioki) , will cause liquefaction at soft deoosites in the coastal regions as shown in

Figure 1

.

Hokuriku Region

Fig. 4 shows the liquefaction history of Prefectures of Yamagata and Niigata during

the past century. Earthquake No. 6 (Shonai) , caused liquefaction at many sites, which

are denoted by hollow triangles as shown in Fig. 4(b). In Sakata City, located at the

river mouth of Mogami River, numerous sand boils were observed with the largest sand

volcano reported having a height of 60 cm and about 3 m in diameter. The area where the

sand boiled most violently was covered by sand deposits, similar to seashore areas. Similar violent sand boils were observed at site 2, shown in Fig. 4(b), where the largest

one was about 9 m in diameter. The liquefaction site is at the same location as caused by

earthquake No. 42 (Niigata) , as shown in Fig. 4 (b)

.

Earthquake NO. 42 (Niigata) is very famous for its wide spread liquefaction, and the

subsequent damage to modern buildings and civil engineering structures. Because the

details of this earthquake have been reported in detail by many others , only a summary

will be given in this paper. Liquefied zones during this earthquake are denoted by hatched

zones, with numbers of 1 to 13, as shown in Fig. 4. It is important to note that these

zones are limited to the original riverbeds of the rivers of Shinanao, Agano etc. For

example, the zone expressed by the character "a" in Fig. 4 (a) , is the original riverbed

of the Anago River which was connected to the Shinano River at the river mouth about three

V-4
hundred years ago. This zone was then reclaimed about fifty years ago by depositing sand.

The national highway No. 7 passes through this zone, and its low bank composed of 1 m of

sand settled down about 1 m and moved laterally 4 m during the earthquake. This was

caused by liquefaction of the sandy bank and sandy ground, because boiled sand covered the

road surface and one truck on the bank sank into the ground.

Discussions and Remarks

The study of the liquefaction phenomena caused by different earthquakes indicated the

following;

(i) Re-liquefaction has observed in 7 zones, shown by symbols "a" to "g" in Fig. 1

and as listed in Table 2. This suggests that re-liquefaction will occur if the

proper soil conditions are present.

(ii) One of the causes of liquefaction in sub-soils is the intensity of the quake.

Earthquake intensity, at the liquefied site, were estimated by comparing the

liquefaction distribution map and the intensity scale distribution map. From this

procedure, it was found that the estimated earthquake intensity in JMA (Japan Meteorological Agency) scale at the liquefied sites was greater than five, which means that

accelerations of 80 gals to 250 gals occur

(iii) As is well known, liquefaction zones increase during the larger earthquakes. Fig.

5 shows the maximum epicentrical distance of the liquefied sites R (km) and the magnitude of the earthquake M. Using the procedure presented by Fukuoka (1971) , the following

is obtained;

log^^R = M-5.7 (1)

where Eq. (1) is based on the data obtained from earthquakes NO. 33 (Nankai) and No.

42 (Niigata) . It should also be noted that examination of Fig. 5 indicates the following

log^^R = 0.87 M-4.5 (2)

where the lower bound for M > 6.0 is expressed by

log^^R = 0.77 M-3.6 (3)

From the data obtained on 44 earthquakes, a lower bound of the liquefaction R-M

relationship can be anticipated by Eq. (3) or Fig. 5. On the basis of this relation

liquefaction may not occur at any site when the distant site exceeds that value

expressed by Eq. (3) . For M = 70, for example, liquefaction may not occur at

distance greater than 60 km from the ^i center.

V-5
Conclusions

A survey of many reports on earthquakes during the last century, shows that liquefaction

in sub-soils are induced at many sites. In Japan during 44 earthquakes, liquefied sites were limited to alluvial deposits and reclaimed lands. Liquefaction generally occurred at

original river beds which was reclaimed during the past several hundred years and in re- claimed lands along seas or lakes. Such liquefaction increases as the earthquakes with

magnitude increase have an offshore epicenter. Furthermore, it could be noted that liquefaction has occurred repeatedly during different earthquakes in several zones. An estimated JMA intensity scale factor at the liquefied sites, was greater than five. A lower

boxind of liquefaction potential can be defined as a relationship between epicentral distance and magnitude of earthquake on the basis of this discussion.

Acknowledgments

The writers would like to express appreciation to Professor T. Mogami, Professor M.

Fukuoka, Dr. T. Okubo, Assistant Professor K. Ishihara and Mr. T. Iwasaki for their

helpful suggestions. This research was performed with an assistance of Mr. S. Yoshida,

Earthquake Engineering Research Section, P. W.R.I. The writers wish to acknowledge his

helpful cooperations.

V-6
References

1. Fukuoka, M. , (1971), "Memories of Earthquakes and Foundations", Bridges and Foundations,

Vol. 5, No. 10. (in Japanese).

2. lida, K. , (1974), "Past Earthquakes and Damages in Nagoya City", Report of Committee

for Disaster Prevention of Nagoya City (in Japanese)

.

3. Ishihara, K. , (1974), "Liquefaction of Subsurface Soils During Earthquakes", Technocrat,

Vol. 7, No. 5.

4. Kishida, H., (1969), "Characteristics of Liquefied Sands During Mino-Owari, Tohnankai

and Fukui Earthquakes", Soils and Foundations, Vol. IX, No. 1.

5. Kuribayashi, E. , Tatsuoka, F. and Yoshida, S., (1974), "Liquefaction History in Japan

During the Last Century", Bull, of Public Works Research Institute, No. 30. (in Japanese)

6. Miyabe, No., (1933), "On the Abnormality of Ground Water During Earthquakes", Earthquakes, Vol. 5. (in Japanese).

V-7
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V-8
Table 2. Zones where re -liquefactions were observed

a, b g correspond to those in Fig. 1.

Zones T71 J- 1 1 _ Earthquakes

a; the narrow zone along the

Omono River in the

northwest of Oomagari city

in Akita Prefecture

No. 8 (Rikuu) and

No. 13 (Ugosen)

b: on the left side at the river

mouth of Mogami river

No. 6 (Shonai) and

No. 42 (Niigata)

c: in the original river course

of Tone river in the

notheast of Tokyo

No. 5 (Tokyo),

No. 7 (Tonekaryu),

No. 15 (Kanto) and

No. 22 (Nishi-Saitama)

d: on the both sides of

Chikuma river near

Nagano city

No. 9 (Kamitakai) and

No. 28 (Nagano)

e; in the southern part of

Fukui plain

No. 4 (Nobi) and

No. 24 (Fukui)

f: in the nothwestern part of

Nobi plain

No. 4 (Nobi) and

No. 12 (Gono)

g: in the southern part of

Nobi plain

No. 31 (Tonankai) and

No. 32 (Mikawa)

V-9
Fig. 1 Liquefaction distribution map (Japan)

V-10
V-lll
Railways <0 Cities

•^^^^^^=5^ Mountain areas or terraced hills

• (Tokyo) A (Tone-karyu) A (Kanto)

o (Nishi-saitam-a)

Fig. 2 Liquefaction distribution map (Kanto region)

V-12
(Mikawa) —

Fig. 3 Liquefaction distribution map (Nobi region)

V-13
Epicenter

(Shonai)

Liquefied

locations

during the Nagaoka

earthquake

Liquefied

location

during the Sekihara

earthquake

Railways O Cities

Mountain areas or terrased hills

A 1-4 (Shonai) jLiquefied locations

l-13jM(Nii9ata)J

Fig. 4 Liquefaction distribution map (Hokuriku region)

V-14
W spn^LuSeirtl

V-15
VIBRATION TEST ON SETTLEMENT

OF SUBMERGED SAND LAYER

by

K. Sawada

Chief, Soil Dynamics Section

Public Works Research Institute

Ministry of Construction

and

Y. Koga

Research Engineer

Soil Dynamics Section

Public Works Research Institute

Ministry of Construction

ABSTRACT

A series of vibration tests on submerged sand layer model, using a large shaker table,

was perfoirmed in order to establish a method which can predict the liquefaction and settlement phenomenon of sandy ground during earthquakes. As a result of these tests, information

concerning the vibration behaviour during the liquefaction phenomenon and subsequent

settlement behaviour was obtained. Examining these test results and assuming that

the vibration and settlement behaviour of a sand layer in the sand container is one-dimensional, it has been found that the test and analysis results agree quite favorably. Therefore, if the shear stress in the ground can be reasonably estimated, the amount of settlement of the sand due to earthquakes can be estimated by this analysis method proposed

herein.

Key Words: Earthquakes; Liquefaction; Sand Layer; Shake Table; Vibration Tests; Void Ratios.

V-16
Preface

The building of a structure on a sandy ground, where liquefaction is likely to occur at

the time of an earthquake, requires the following sequence of steps in order to perform a aseismic design.

» (1) Determine whether or not a liquefaction phenomenon will occur in the ground which is

to support the building and then establish the degree of reduction of the bearing capacity and the amount of settlement which may be generated. Then estimate whether or not

certain measures are required to minimize the effects.

(2) If design revisions are required, then improvements to the foundation or the type of

structure, or improvements to the ground are necessary so that a liquefaction phenomenon

be either prevented or reduced.

The methods of determining or estimating liquefaction phenomena are based on a method

from sand element tests (1-3)

.

At a previous meeting, a method emptoying blasting vibrations to predict liquefaction,

was presented. The results to be presented herein, however , will involve the analysis of

experiments with a model sand layer using a large shake table. This experiment was performed jointly by the Public Works Research Institute, Ministry of Construction, the Kanto

Engineering and Mechanical Laboratory, Ministry of Construction and the National Research

Center for Disaster Prevention, Science and Technology Agency.

Method of Experiment

The container used for the experiment is a steel box having inner dimensions of 8 m

(length) X 5 m (height) x 2 m (depth) as shown in Photo 1. In the experiment, the box was

installed with its longitudinal direction aligned with the horizontal vibrating direction

of the slab table and the sand layer of thickness of 4 m, was submerged in water up to the

surface. The sand used in the experiment is mountain sand taken from Sengen-Yama in Chiba

Prefecture, Japan, which has physical properties as shown in Table 1. A belt conveyor and

bucket were used to place the sand into the box for forming the sand layer. The sand was compacted by stepping or a vibro-rammer for every sand layer thickness of 30 or 50 cm. A

quality control test, including measurement of density, was performed for each of sand

layer. Upon formation of each sand layer, the layer was then submerged in water. After

inducing series of vibrations and subsequent measurements , the water and sand were discharged from the sand box and a new sand layer was formed for the next experiment. This

type of experiment was conducted three times with respect to a homogenous sand layer.

Three additional tests were conducted with respect to a case where a model object was buried in the sand layer.

The following types of vibrating methods were used in each experiment;

(1) Resonance Test

This test had a table acceleration equal to 20 gal, and a step sine frequency of 1 to

20 c/s. The vibrations were applied for a fixed period of 20 to 30 sec. This test was

intended to examine the dynamic properties, including modulus of elasticity and damping

factor.

V-17
(2) Liquefaction Test

The test level of acceleration of the table was changed by means of sine waves near the

1st resonant frequency and were applied for a fixed period of 60 to 100 sec. This experiment was intended to examine the effects of excess pore water pressure and the resulting

settlement in the sand layer. In this experiment, where the excess pore water pressure was

developed, the sand layer was allowed to stand for about 1 hour after completion of the

vibration test sequence. The application of next vibration stage was initiated after the excess pore water pressure was dissipated.

Table 2 describes the conditions for each model experiment, including the number of

times the sand layer has formed the vibrating conditions.

Illustrated in Fig. 1 is an example of the arrangement of the instruments buried in

the sand layer.

Results of Experiment

(1) Characteristics of the Sand Layer

The density of sand layer varied considerably with each model, as affected by the

method of compaction. For example, in the case of the 1st experiment, the dry density (y,)

3 3 ^ was 1.41~1.52 g/cm , the average value was 1.48 g/cm , void ratio (e) was 0.77~0.91

and the average value was about 0 . 82

.

(2) Resonance Test

Examples of the resonance curve obtained from resonance tests are illustrated in

Fig. 2(a) through (c) . Illustrated in Fig. 2(a) and (b) are the response characteristics

of the sand layer for experiment 1-1, relative to distance from the end face. For this

experiment, the frequencies at the 1st and 2nd resonances are 10 and 17 c/s respectively.

This shows that there is little difference relative to distance from the end face, but it

can be seen that the magnification of acceleration varies with the distance from the end

face. For example, with respect to 10 c/s which can be thought of a 1st reso'nance frequency

Fig. 3 shows the distribution of the magnification of the acceleration which indicates that

the farther from the end face, the larger the magnification of the acceleration. Experiment

3-1, as illustrated in Fig. 2(e), is different from the experiment 1-1 as illustrated in

Fig. 2(a) and (b) with respect to density and table acceleration level.

(3) Liquefaction Test

The characteristics of the behaviour of the sand layer during liquefaction can be

represented by the pore water pressure, settlement amount and acceleration of the sand layer.

Examples of the time-course changes of these factors are illustrated in Fig. 4 through

Fig. 6. Illustrated in Fig. 4 is an example in which both the excess pore water pressure

and the settlement are small due to the small table acceleration. In this case, the

acceleration of the sand layer except the upper part of the sand layer shows almost a steady

state where the table acceleration is constant. The acceleration of the upper part of the

sand layer shows a trend in which it continues to increase even after the table acceleration

has become steady and then suddenly decreases after it has reached a peak value. This

phenonmena is caused by first an increase of the excess pore water pressure in the upper

V-18
part of the sand layer. This increase in the excess pore water pressure over a certain value

causes the rigidity of the sand layer to lower (or causes the deformation characteristics to

increase) due to a reduction of the effective confining pressure. This resulting phenomenon

causes an increase in acceleration of the sand layer, and if the excess pore water pressure

further increases, the shear strength of that layer and the lower layer will reduce. This

will result in a condition in that the vibration from the lower part cannot be transmitted

to the upper part and the acceleration of the sand layer will suddenly decrease.

In the case of experiment No. 1-4 (Fig. 5) , the table acceleration is larger than

that of experiment No. 1-3 (Fig. 4) , and both the excess pore water pressure and the settlement amount are comparatively larger than those in the experiment No. 1-3.

In experiment No. 1-4, the excess pore water pressure increases starting from the

upper part of the sand layer, with a maximum value nearly equalto the effective overburden

pressure in each layer. This conditions shows that all layers have reached complete

liquefaction. The range in which the phenomenon of a sudden decrease in acceleration of the

sand layer occurs is deeper than that sand layer in the experiment No. 1-3, and researched

a depth of 3 meters.

Experiment No. 3-4 (Fig. 6) , is the case where the void ratio of sand layer is less

than that of experiment No. 1-4. However, despite the fact that the table acceleration

and vibrating period was larger than those generated experiment No. 1-4, the excess pore

water pressure and the amount of settlement generated was smaller.

Fig. 7 shows the vertical distribution of the acceleration and excess pore water pressure previously given by Fig. 5. The process in which a sudden decrease of acceleration on

the upper part of the sand layer progresses to a deeper part with the passing of time can readily be seen. One thing of interest in comparing the verical distribution of excess

pore water pressure, as illustrated in (b) and the corresponding experiment period versus

the amount of settlement (Fig. 5(c)) , is that for the period between 0 to 40 sec, the water

pressure gradient at a position 1 m from the bottom is downward and the pore water tends to

flow into the lower layer, so that at the position 1 m from the bottom, uplift must occur

rather than settlement, as illustrated in Fig. 5(c), during the period, a considerably

amount of settlement is observed. This condition is caused by air bubbles in the pores

which have been compressed due to the generation of excess pore water pressure, since the

sand layer had not been completely saturated. This mechanism can also be concerned by noticing first that the water level, on the sand layer surface, after vibration is lower by

several cm compared to the level before vibration is applied. In addition, since the degree of saturation of the pore water of the sand layer can not be measured directly, an unsaturated sand layer was formed separately in a small container, similar to that of the

experimental sand layer. Then the unsaturated sand layer was gradually submerged in water

from the bottom. The initial degree of saturation of the sand layers formed by the method,

was then estimated at 75 to 85% by the weight measurement method and by means of an air

meter used to measure the air volume in concrete.

V-19
Discussion

As described earlier, the primary aim of the resonance test is to calculate the vibration behaviour of the sand layer subjected to larg<; dnput vibrations, and to compare the

results with analytical values. In this paper, however the analytical results will be

omitted and we will discuss the method of analyzing the liquefaction and settlement processes of the experimental sand layer.

(1) One-dimensional settlement of the sand layer

The vibration behavior of the sand layer in the sand box, and the subsequent deformations are three-dimensional. However, since the excess pore water pressure, that is

measured shows almost the same behavior regardless of the distance from the end face at the

same depth, the amount of settlement can be presumed to ,be one-dimensional. Furthermore, it

is generally difficult to analyze multidimensional deformations, therefore, if a one-dimensional settlements can be estimated, the results can be adequately applied to actual problems. Thus for the experimental sand layer, such assumptions were made considering the

horizontal ground expanding indefinitely and then analyzing the process of liquefaction and

settlement.

(a) In thenormal theory of consolidation, the deformations of a sand layer are generated

as a result of the outflow of pore water. In these tests, however the sand layer

is assumed to be in an unsaturated state with air remaining in the pores and

that the air is combined with the pore water and flows together. This assumes that

the porewater is compressible.

(b) It is thought that the excess pore water pressure, generating a pore water flow, is

generated internally by cyclic shear.

(c) It is assumed that the ordinary Darey's law can be applied to the pore water flow.

From these assumptions , the excess pore water pressure and the void ratio in the sand

layer can be represented as follows;

Ju = B./(-^ e, Ko)-<Jvo* T C d

_r d (1) vc

je = (1-B) (l+e)-C./(-7r-. e, K^) •

• ^ vc

+ B (1+e) • k 9 u

r dZ' ' 0) At (2)

V-20
where,

B

C

Pore pressure coefficient l+eC^/(l+e)C (3>

Compressibility of sand

Compressibility of pore water (including air)

1-(1-H)S

r 1

H

S

Pl

f

"^d a

' vc

eK

o

a

' vo

T

k

Y.,

pl + Au

Henry's coefficient of solubility

Degree of saturation

Absolute pore pressure (added value by atmospheric pressure and static water

pressure, considered to be equal to pore water pressure and pore air pressure)

Pore water pressure generation function due to cyclic shear (described later)

Cyclic shear stress

Effective overburden pressure used for pore water pressure generation function

Void ratio

Coefficient of horizontal earth pressure at rest

Initial effective overburden pressure

A period of vibration

Coefficient of permeability

Unit weight of water

Depth from the surface of sand layer

The meaning of the pore water pressure generation function f by cyclic shear shall b

defined as follows. In the instance of saturated and undrained conditions, if one cycle i

applied, the following equation can be obtained, where C^=0, B=l, k=0 and t=T, in Eq. (1)

Ju = f {^-^, e. Ko)-<vo

'VC

(5)

J e

Consequently, f*a'^^ represents the pore water pressure generated per cycle under said

conditions. This is based on results of undrained, cyclic, simple shear tests which have

been reported. This pore water pressure generation function f was obtained based on the

following assumptions and method.

(a) The increase in the excess pore water pressure, sand subjected to cyclic shear,

is the sum of the increments due to cyclic shear and the increments due to an in4) crease in lateral pressure due to the lateral displacement being confined

V-21
Consequently, the excess pore water pressure Au generated per cycle of cyclic

shear the sand is given by the following equation;

<^'vo ^1 S + 2 Ko ' <^'vo ' (6)

7— n i - v , ^ • —:7T

Hence, f^ can be obtained by the isotropic cyclic triaxial test as follows;

J u , (J d

^1 ^ ^30 ^ :i " 30

0*3 0 : The initial effective lateral pressure

CJ, : Cyclic axial stress

d

(b) The pore water pressure generation function f^ is obtained from a conventional

cyclic triaxial test. These tests yield the stress and the generated pore

water pressure. However, there is generally a wide variance in the test results.

Consequently, by plotting the relationship of "^d/2'^' 3ovsn„) obtianed from the lique5) faction test, an average relation is estimated from the following equation ;

id _1

Where o^: Angle of internal friction mobilized when the initial liquefaction is

generated,

n^: Cycles of loading for generating the initial liquefaction

For the tests reported herein, C?d/2a'3o n^^^ is plotted as shown in Fig. 8.

(c) A relationship for general density can be obtained from the following equation

with respect to the stress ratio at which liquefaction is generated;

''d , , V Dp

) ^ = ( . ,r ) 2.''o ' D„ ~ 2^' 'Dp=50% -50- (9)

Based on these assumptions, the function of the generation of pore water pressure in

sand is estimated as illustrated in Fig. 9. This function is used as f in the previous

equation (1) and (2)

.

The increments A u and A e, the excess pore water pressure and void ratio re- spectively from equation (1) and (2) , are closely related to each other as shown in Fig. 10.

This diagram can be interpreted as follows, i.e. path AB represents the entire process that

V-22 i
occurs, which means first reduction of the effective stress due to an excess pore water pressure generated when the sand layer is vibrated. The pore water then flows out and as a

result the void ratio is changed. Then the excess pore water pressure is entirely dessipated and^ stability is attained. Path CF represents the changed path which occurs in a short

time interval At. If this path is divided into CF = CE + EF, CE corresponds to the 1st

terms of equations (1) and (2) respectively. Thus the 2nd term

deformation path in this case of the undrained condition. An alternate path occurs when the

sand layer is placed under a drained condition as shown in path EF. That is the deformation

under the drained condition of an unsaturated sand layer in a short time interval can be

obtained as the sum of the deformation under the undrained condition of an unsaturated sand

and the deformation under the drained condition given thereafter.

The coefficient of compressibility C in equations (1) and (2) was obtained in the

following manner. As shown for path EF, the excess pore water pressure was generated by

the undrained cyclic shear and then by draining of the sand. The coefficient of compressibility in the recompression process is then obtained. An example of this experiment is

illustrated in Fig. 11.

Also illustrated in the same diagram is the static virgin compression curve of this

sand. Examining these results the following can be stated;

(a) Until the initial liquefaction point is exceeded, the coefficient of compressibility

in the recompression process is smaller than the coefficient of compressibility under the

virgin compression process.

(b) When the sand is recompressed, exceeding the initial liquefaction point, the compressibility is as nearly same as the virgin compressibility curve.

(2) Numerical analysis of liquefaction and settlement process

Considering the above, the following assumptions were made in the numerical analyses of

the liquefaction and settlement processes as will be described later.

(a) When an excess pore water pressure is generated and exceeds the initial liquefaction

point, thus causing compression of the sand, a virgin compression curve is used.

(b) During the process, in which the density increases higher than that of the past

minimum void ratio, a virgin compression curve is used.

(c) In the deformation paths other than the above (a) and (b) , an unloading or recom- pression curve is used.

(d) The generation of an excess pore water pressure function f does not occur until the

time the excess pore water pressure reaches the initial liquefaction point, CJ'^^ =

when the initial liquefaction point has been exceeded, O^^ = a^' (0^' represents the

effective overburden pressure at that time) , and the working shear stress also decreases in

the form of contacting with the failure envelope line.

Since a vibration analysis and subsequent evaluation of the shear stress of a sand layer,

which has changes in its physical properties subsequent to an increase in the excess pore

water pressure, is difficult to determine, the following method was employed.

V-23
(a) The vibration mode of the sand layer, for liquefaction test, was assumed as the 1st

mode. The displacement amplitude was obtained from the amplitude of the measured acceleration and the distribution of the shear stress is then estimated from these results. During

this time, when the table acceleration level increases, as seen in Fig. 5(a) on the upper

part of the sand layer, the acceleration decreases due to generation of the excess pore

water pressure thus making it difficult to estimate the shear strain. Therefore, assuming

that the shear strength does not decrease, the standard shear strain distribution corres- ponding to the individual table acceleration level is estimated from this study.

(b) Now by multiplying the standard shear strain distribution by the shear modulus, in

7) which the reduction effect of the strain level is introduced from Richart's experimental

6

)

equation for the shear modulus in an infinitesimal strain level of sand, a shear stress

distribution is calculated. When the excess pore water pressure exceeds the initial liquefaction point, this shear stress is decreased to a form which contacts the failure envelope.

Examples of changes of the excess pore water pressure and the settlement amount with time

have been calculated for the experiments and are illustrated in Figs. 12 and 13. Figs. 12

and 13 correspond to experiment No. 1-3, experiment No. 1-4 as illustrated in Fig. 4 and 5. According to the calculated results for the experiment No. 1-3, the excess pore water

pressures in the upper layer (Z = 0.5, 1.0 m) increases considerably, consequently in the

earlier stage of vibration the settlement amount is a little larger than that of the

experiment. Also the excess pore water pressure is larger than measured in middle part of

the sand layer and the final calculated settlement amount is larger than that measured.

According to the calculated values for experiment No. 1-4, the excess pore pressures in the

lower layer (Z = 3.0, 4.0 m) increases slowly, but the settlement behavior agrees approximately with the measured results. The cause for these differences, apart from the assumptions in the calculations, are the large errors in estimating the shear strain in the sand

layer and compaction due to the preceding vibration. The settlement behaviour of the sand

layer shows that the calculated results agree comparatively well with the measured values.

From the aforementioned, the method for liquefaction and settlement analysis introduced

herein is rather effective to the analysis for this vibration test of model sand layer.

Conclusion

The following conclusions can be derived from the results of the experiments and analyses as described previously;

(1) In the liequfaction test, when the excess pore water pressure increases to a certain value, the magnification of the acceleration increases and then suddenly decreases.

(2) The higher the table acceleration and the larger the void ratio, the larger is the excess pore water pressure and the amount of settlement that is generated.

(3) The compression phenomenon of the air in the pores increases the settlement of the

sand layer.

(4) From examination of the experimental results in line with the analytical method proposed

herein, it has been found that there is excellent correspondence between the experiment

V-24
data and the analysis.

If this method is used to estimate the amount of settlement of the sand at a construction site, it is important to properly estimate the shear stress in the sand. Also, at the

construction site, the ground below the water level is saturated and the duration

of an earthquake is relatively shorter than that which occured in the experiment described

this paper. Consequently, a simplified method, has been developed. This method is based

on the assumption that the amount of settlement during an earthquake is small, and therefore the ground is nearly in an undrained condition. The excess pore water pressure then

generated is estimated, and the settlement amount is determined as a static deformation due

to the outflow of the pore water after the earthquake.

V-25
References

1. Seed, H.B. and Idriss, I.M.: Simplified Procedure for Evaluating Soil Liquefaction

Potential, Proc. of ASCE, SM 9, 1971.

2. Japan Road Association: Aseismic Design Mannual of Road Bridge (in Japanese), 1972.

3. Florin, V.A. and Ivanov, P.L.: Liquefaction of Saturated Sandy Soils, Proc. of 5th

Int. Conf. on S.M.&F.E., Vol. 1, 1961,

4. Ishihara, K. and Li, S. : Liquefaction of Saturated Sand in Triaxial Torsion Shear

Test, Soils and Foundation, Vol. 12, No. 2, 1972.

5. Shibata, T. : Liquefaction Phenomenon of Saturated Sand (in Japanese), Preprints of

16th Symposium of SM (JSSMFE) , 1971.

6. Richart, F.E., Jr., Hall, J.R. and Wood, R.D. : Vibrations of Soils and Foundations,

Prentice-Hall, Inc., 1970.

7. Hardin, B.O. and Drnvich, V.P. : Shear Modulus and Damping in Soils: Design Equations

and Curves, Proc. of ASCE, SM 7, 1972.

V-26
Table 1 Index properties of the test sand

Properties

Specific gravity Gs 2. 71

Optimum water content ^opt (%) 13.4

Maximum dry density rd max (g/cm^) 1. 742

Maximum void ratio e max 1. 02

Minimum void ratio e min 0. 57

100

I 80

c 60

0.01 0.1 1.0 10.0

Particle diameter (mm)

Table 2 Test Condition

Vibration

Method

Model

Resonance

test

Liquefaction test

1 2 3 4 5

1

6 ~ 13 gal

1 ~ 20 c/s

eo = 0. 82

37 *gal

12 c/s

74**sec

61 gal

11 c/s

76 sec

138 gal

10 c/s

84 sec

274 gal

9 c/s

92 sec

2

eo = 0, 73

57 gal

12 c/s

69 sec

100 gal

11 c/s

72 sec

132 gal

10 c/s

81 sec

283 gal

9 c/s

78 sec

3

7-19 gal

1 ~ 20 c/s

eo = 0. 68

66 gal

12 c/s

93 sec

99 gal

11 c/s

88 sec

219 gal

10 c/s

103 sec

365 gal

9 c/s

85 sec

* Stationary value of table acc. ** Total duration of vibration

V-27
Photo. 1 Model sand box

4m A A A

1- A A

o o o O

3 0 A AO A i 0 O

2 o Za AO AO 1 A AOf

1 0 A AO AO 0 o

0

0 A A AO 0 o

1 2 3 4 5 6 7 Sm

1 1

\ 1 I I I J

A Accelerometer o Pore pressure gage

i Settlement gage

Fig, 1 Arrangement of meters

V-28
IS

Act of Shaking Ubic [ 10 2

(gal)

:| I3.2i 11.4 1 1 1 : I 107 1 Q ? I 8.1 I 6 6 I 9 7 I bb I Q.O I 8.1 1 lO.Ol i:-S i M.0| B.3 I 9.0 I 10.3| 10

I 8.6 I 9*4I 9.7 j 9.4 I 10.7 j IQ.OI

A-n/A-31

A-14/A-3I A- 9/A-3I

A- 6/A-3I A- 3/A-31

jSing laWc |l.i<l n7|ia.5|l7.0|l7.0|ia,Sll8-S|l8.S|l7.7|lS.4|n0|lS4|l6.:|l7.9|lS.4|l5 0ll3.Q|i:..M -m'TT

(^1)

"

I

I V4! 15.4' IS.4I 15 4 ; lb 7 I l6-2> 1S.4- lfi.:|

Fig. 2 Examples of resonance curves

V-29
Acceleration ratio

0 2 4 6 8 0

TEST NO 1 • 1

FREQ 10 CPS

8 0 8 0 8 0

8m

350

Fig. 3 Distribution of acceleration in sand layer

Fig. 4 Acceleration, excess pore water pressure and

and settlement data (Test No. 1-3)

v-30
8

-

Fig. 5 Acceleration, excess pore water pressure and

settlement data (Test No. 1-4)

v-31
500

Fig. 6 Acceleration, excess pore water pressure

and settlement data (Test No, 3-4)

v-32
Fig. 7 Vertical distribution of acceleration and excess pore

water pressure

0 I I 1

1—

I 10 100 1000

Cycles to liquefaction n|;

Fig. 8 Liquefaction characteristics of test sand

V-33
o->

to "a<

O.IO

0.05

Dr: Relative density ^

Ko= 0.5 .<=)

0.15 0.20 0.25

Stress ratio rd/ovo

Fig. 9 Generation function of

excess pore water

pressure

EfTective streg p

Fig. 10 Schematic relationship

between void ratio and

effective stress due to

vibration

EfTective stress p (kg/cm')

Fig. 11 Compression characteristics of

test sand

V-34
Fig. 12 Calculated results of excess pore water pressure and

settlement (Test No. 1-3)

- 200

8

•

Fig. 13 Calculated results of excess pore water pressure

and settlement (Test No. 1-4)

v-35

STUDY ON EARTHQUAKE RESPONSE OF

STRUCTURES BY CONSIDERING NON- DETERMINISTIC

VARIABLES

by

Yutaki Yamazaki

Structures Division

and

Yasunori Koizumi

Director

Building Research Institute

Ministry of Construction

ABSTRACT

The seismic design coefficient has been widely used because of its simplicity. This

design technique provides a safeguard against earthquakes but contains many problems when

replacing the dynamic forces with static forces. In addition, the earthquake resistant properties of high-rises or buildings can be investigated by a simulation analysis, using many

past earthquake records as input excitation. This type of dynamic analysis of a structure

has been made possible by use of electronic computers. In this analysis, an actual structure is transformed into a vibration model, then the structure is subjected to earthquake

ground motions, which are simulated by an electronic computer. As an alternate scheme,

from a stochastic point of view, vibrational properties of structures subjected to earthquake

ground motions have been investigated by utilizing the concept of random vibrations. The

theory of random vibration, for dynanic response of structures, considers that earthquake

ground motions can be essentially predicted as deterministic phenomena and that the vibrational behaviour of structures during these earthquakes can be predicted stochastically.

Studies on developing a more reasonable design method have been conducted using the experiences obtained from earthquake disasters and the analyses of earthquake phenomena. The nondeterministic phenomena of earthquake ground motions will be treated essentially by the

application of the theory of random vibration to the earthquake engineering. However, it

is also true that a structure cannot be handled as a deterministic system, because the

dynamic property variables of an actual structure, such as the masses, spring constants and

damping constants , cannot be evaluated deterministically when the structure is desinged.

Hence, a structure must be designed by considering the nondeterministic properties of the

structure as well as those of earthquake ground motions. In this thesis, a theoretical

treatment of the earthquake response problems of a structure with nondeterministic variables

have been discussed and a reasonable design technique of a structure with appropriate safety

has been suggested.

Key Words: Probability, deterministic, earthquake, structures, random vibrations,

safety, dynamics.

VI-1
Introduction

Vibrational properties of structures, subjected to earthquake ground motions, have

been investigated by utilizing the concept of random vibrations. The theory of random

vibrations for the dynamic responses of structures is based on the stochastic point of

view that earthquake ground motions cannot be predicted as a deterministic phenomena and

that the vibrational behaviour of structures during earthquakes must be investigated stochastically. However, it is also true that a structure cannot be examined as a deterministic system, because the dynamic property variables of an actual structure, such as masses, spring constants and damping constants, cannot be evaluated deterministically when

the structure is designed. Hence, a structure must be designed by considering the non- deterministic properties of the structure as well as those of the earthquake ground motions

There has been many papers concerning the dynamic response of a structure subjected to

earthquake ground motions by application of the theory of random vibrations since E.

Rosenblueth and H. Tajimi^^^. The theory for a stationary random response of a linear

(3) lumped-mass system has been studied by one of the authors . Expansion of this theory

is explained in this paper, in which all variables are regarded as non-deterministic. The

fundamental formula obtained by Taylor's series of a function f(r), of the vector of ran- dom variables r, is applied to this expansion. There are papers ^^^in which the properties of stochastic variation of the eigen values and earthquake responses of a structure

are investigated which demonstrate application of this formula. In this paper, the theoret

ical expansion is conducted in order to obtain a general and meaningful solution of the

earthquake response of a structure. The theory of the earthquake response of a structure

described herein has the following characteristics;

(1) The structure is a one-dimensional linear multi-degree-of-freedom system with

masses, springs and dashpots.

(2) The structure is subjected to a stationary random excitation, having any distribution of power spectral density

(3) The structural response can be stochastically predicted in consideration of the

interaction effect between the structure and the ground.

Response of a Structure Subjected to a Random Disturbance

The earthquake response of one-dimensional linear multi-degree-of-freedom system with

masses, springs, and dashpots as shown in Fig. 1 is analyzed as follows; The interaction

effect between a structure and the ground is considered by application of sway and rocking

springs. The earthquake ground motion, which affects the structure, is assumed to be a

stationary random process with an amplitude distribution of Gaussian order with a zero mean value.

The equations of motion of the system subjected to an earthquake ground acceleration

xg is generally represented by

{Z ) + [C J iZ 1 + fK] { Z I = -{MeI

(1)
where

w =

(Me) =

(Z) =

O

mo

(2.)

{m, nij

{Z, Zz Z M Z 0 0

fK] =

r ~ ~

1 -k, 0 0 0 • • • 0 1 0 -k.fl:

1 -k. k, + k2 -k2 0 j 0 k.Hl -kill

1 0 -kj kz+ka 0 kzHl-kaiTT

1 ^ 0 -kj

0 1

-kN-1

i

0

0

0

kiHi-ki+iHi+i

kN-2H N-2- kN-l Hn- I

1 ^ 0 0 kN-i+kf, 1 -kN kN-i Hn-1 - kNHN

0 0 0 0 -kN khftlcs kxHN

kiHl-kjITz kziTz-ksHl kN- Hn-1 -knHN kNHN

N _2

kR+I kiHi

i=i /

(3)

m, mo and I: masses of each story and the foundation and the moment of inertia about

the center of the foundation, respectively

k, ks and kR: spring constants of each story, sway and rocking springs, respectively

H: story height

The damping matrix [C] is prescribed here by

where (jO and h is the fundamental circular frequency and damping constant of the system,

1 1 respectively.

VI-3
The vector of the relative floor displacement {z} is expressed in terms of the normal

mode matrix [(})] and the vector of normal coordinate {r|}, namely

{z} = (4)){n} (5)

Substituting Eq. 5 into Eq. 1, the equations of motion of the system are expressed into

each mode as follows.

?j ' 2CJ '^j hi + 1, = "'^J (6)

where

. (7)

(8)

In the above expression, the mode matrix [cj)] is normalized. In this case, a generalized

mass matrix M* , given by Eq. 9, corresponds to a unit matrix E

The autocorrelation function of the normal coordinate Tlj is represented by

where S^jj (p) is the power spectral density function of rij • Defining H_. (ip) and S^^Cp) as

the function of the frequency response of the system and the power spectral density function

of a ground acceleration xg, respectively, it follows that:

S7j(P) = |Hj(iP) r S'^g(P)-/J' (11)

R'7j(0=r/j"" |Hi(iP)r e*"^ S'', (P) dp

(12)

Utilizing the relation

the autocorrelation function of is represented by

R7JjCr) |Hj(iP)r P'e'P- S';(P)dp (13)

VI-4
Representing the power spectral density function of the ground acceleration xg as a sum of

the power spectral density functions of the single-degree-of-freedom systems excited by

white noise accelerations, the function is expressed by

S = l

where LO , ^ and s are the natural circular frequency and the damping constant of the singledegree-of- freedom system and the constant power spectral density of the white noise acceleration, respectively. In general, the following relationship exists between the response

Z. of i-th story and Z of the s-th one.

1 s

Zi Zs — S.S <t>ij <i>tv ?j'?k = .?^ij^8j v] = ^ <t>i) R^j (0) (15)

J K J j

From the above relationship, the mean square values of the responses Z^ and Z^ become

(16)

Z? =J R7j (0)

Zf =S<1,\-^ R^j (O)

^ (17)

Substituting Eqs. 12 to 14 into Eqs. 16 and 17, the following equations are finally obtained.

Z^ = I J <t>\i x,i ^"

(18)

where

C8j = 2Aaj 3Asj-iAsj Asj+aAajeAl

D8j = iA5j ( 2Asj sAsj-iAaj Asj )-3A,j^ (19)

Es j = 3 Asj + I Asj 6 A|

, Asj = sAj -r-eAa and

J

2 Asj + '^'^'^

^ _ aAj jA^

'Asj - ,^^2 ,.2

1 (20)

VI-5
« A, —

0>j

(20)

The mean square values of the relative displacement and velocity at each story are given

by similar expressions namely;

(z,-z.+i)' = j^Uu-*'+")' -oir (21)

The mean square values of the relative story displacement and velocity, substracting the

rocking effect, are also given by

(Z,-Z, + i-H,ZN+2)' = -y^ ( 0ij-<>i + iJ-H,<»N+2j)' '.J

(22)

(Z,-Z, + i -HiZn+2)' =^^J *lJ-<*i + U-Hl^N+2jP z.j

The mean squares of the responses are given by Eqs. 18, 21 and 22. The maximum displacements can be estimated by application of Eq. 23, derived by S.O. Rice, to these mean

squares, providing the process is a stationary random and a Gaussian distribution with zero mean value.

where 0^, and T represent the standard deviation of the responses x and x and the

duration time of the responses, respectively. The mean squares, obtained by Eqs. 18, 21

and 22, correspond to the standard deviation of the responses which have zero mean values.

Hence, the maximum responses can be estimated by substituting Eqs. 18, 21 and 22 into Eq.

23, respectively.

Mean Value and Variance of a Function Composed of Stochastic Variables

Considering f (r) being a function of the vector of random variables r, this function

can be expanded in Taylor's series about the mean values of r. Truncating the series after

the third term, the following equations are obtained.

f ( r ) = f (7) +^ ( r,-7, ) + I

C r,-r, > ^3^ ^ ^) (24)

VI-6
Based on the assumption that the variables r are stochastically independent of each other,

namely the covariance of r^^ and r^^ is equal to zero for the case of k;^?., the mean value of

the function f (r) is directly obtained from this equation , which gives;

f(r)=ECf(r)]=Eff(r)]+|E [(S(r,-r,) "^^'fCr)]

= T^^'^'^'e^ (25)

where Var (r^) means the variance of the variable r^. The variance of the function,

Var(f(r)), is given by

Var (f{r)) = E[(f(r) - f(r)) ]

Substituting Eqs. 24 and 25 into the above equation, the following relation is finally

obtained;

Var(f(r))=^ f^^)' Var ( r, ) + j X X (^^I^)' Var( r, )Var ( r, ) e dre Ike dr^^dti (26)

Eqs. 25 and 26 are the fundamental formulas concerning the mean value and variance of the

function f{r), respectively. Neglecting the third term of the right hand side of Eq. 24,

the following formulas are obtained, instead of Eqs. 25 and 26;

(27)

f (r) = f (r)

Var(f(r))=^, (-777"-^ "^"^"^'^

(28)

The pair of Eqs. 25 - 26 and Eqs. 27 - 28 are called the formulas of " the Second Expansion

of Probability Variance " and " the first Expansion of Probability Variance ", respectively,

in this paper.

Earthquake Response of Structures by Considering Nondeterministic Variables

The theoretical approach, concerning an earthquake response of a structure subjected

to a random disturbance, has been given in part 2. The fundamental stochastic variables r, considered in this application, are masses m, spring constants k, a fundamental damping

constant of the parameters concerning power spectral properties of the ground motions,

namely O) , E and s. All the variables are shown in Table 1. Consequently, the number of

g

fundamental stochastic variables are 2N+3L+6, where N and L are the number of stories of

the structure and the number of peaks on the spectrum of the ground motions, respectively.

VI-7
Applying the formulas of "the First Expansion" Eqs. 27 and 28 into the maximum response

|x|max given by Eq. 23, the following equations are obtained.

(r) = |xL„(7) ^29)

Var(|xL.,(r))=X^(-ii4^/ Var(r,)

^3^^

The former equation can be directly evaluated. However in order to solve the letter equation,

the following is required. First Differentiate Eq. 23 with respect to variable r^^, which

gives the following equations for each condition of the subscript I.

for £=1 (r^^ corresponds to duration time T)

5|xL„(r) _ ol 1 ^^^^

dre T ^- - ' T

2o\ log. ( -

K

for £^1

(32)

2 2 ~r '2 The variance and correspond to mean squares Z and Z , respectively, for relative

displacements to the ground, namely

ti Z j

(33)

The following equations are directly obtained by differentiating Eqs. 33 and 18 with respect

to r^.

VI-8
and

~^ = Z S

ore * i

SX.i N C.j

(35)

d re • Jt

2<5u (

\ E,j / D.j

For relative story displacements and those substracting the rocking effect, Eqs. 21 and 22

are applied instead of Eq. 18. The unknowns in the above equations are differential of

X./C.,D.,E., and (b . . . These are derived in the following

'^s: S3 s] s: ^i:

8x . S3 .

Differentiating Eq. 18-1 with respect to r^ , the differential of X is represented by

A'-

d

where 9rj/3r£ is obtained by Eq. 8, or;

drg dri dre

dre ' 9re

(37)

The differential of C ., D ., and E ., refering to Eqs. 19-2 to 19-4 are represented as

S3 S3 S3

follows;

VI-9
-iKi (-

dre •

- dr^— + 2jA,j «A. I, )

(38)

^= ( 3A.J - 2.A. + A., ) (^^) +.A., ,A^ ( a, A.J ^ , . . c ^'A.j

^

dre

drg

+ ( iA.j 2A.J - 2,A.j ) C g^/ ) - ' A.J g

^ d«A,j

(39)

( ^-^.i) + (^AlL) + 2.A.J .A. (-4^7^)

" • ^ 9re ^ ^ dre

(40)

The unknowns, 9.A /3rp in the above equations are derived from Eqs. 20-1 to 20-6,

respectively;

^1 A,j a re

dre

diAsi

dre

d*A,j

dre

d,Aj

dre

goA.

dre

(o) dre ^^e <•>£, dre ^^e

<o] (ol, dre <"j dre

2 (5) Wj tfr^ &)] dre

(41)

VI-10
where '^^./dr^ is obtained Eq. 7, namely

"a77 " 0,, ^ ^ o,, ^ ^ o,,' dre (42)

The other unknowns in the above equations, namely "b^ . ./ Br^, and 9to./9r„, are obtained by an

(4) (5) application of the formulas of "the First Expansion" to an eigen value problem.

Therefore, the right hand side in Eq. 3- can be calculated with the above equations, which

are represented by the differential of the fundamental variables shown in Table 1. Therefore, the mean value and variance of the maximum response |x|max of a structure can be

theoretically estimated by Eqs. 29 and 30, if the mean value and variance of the fundamental

variables are determined. The above-mentioned relationship, used to obtain the mean value

and variance of the maximum response of a structure subjected to earthquake ground motions,

are based on the formulas of "the First Expansion", namely Eqs. 27 and 28. Applying the

formulas of "the Second Expansion" represented by Eqs. 25 and 26 instead of "the First

Expansion" to this problem, instead of Eqs. 29 and 30, the following equations are obtained;

(43)

,(r)- |xL„(7) 4-i^^ |xL„(T))Var(r,)

Var( |xL„(r) )=^^ (-977 1 x L„(7) )' Var(r,)

(44)

The remaining unknown term is -r— (Ixl (r) ) , in the above equations as the first

dr^dr^ ' 'max

terms of the right hand side in the above both equations have been derived. This term can

be easily derived in the same was as described above. The derivation of the theoretical

relationship concerning "the Second Expansion" is omitted in this paper.

Earthquake Response of a Three-Storied Structure

In this section, the theory concerning the earthquake response of a structure, described

in the previous section, is applied to a 3-story reinforced concrete structure with sway

and rocking springs under the foundation. The dynamic behaviour of the structure shown in

Fig. 2 is analyzed as follows. The constants of the structure, such as masses and spring

constants, are shown in Table 2, where m^, I, k^ and k^^ are the mass of the foundation, the

moment of inertia about the center of the foundation, sway and rocking spring constants,

respectively. The sway and rocking spring constants, k and k , are obtained by the

VI-11
relationship, in which k and k are the dynamic coefficients of subgrade reaction in H V

horizontal and vertical directions, respectively.

BL ^

B and L: breadth and longitude of the foundation, respectively

3 k = klg.cm

H k^ = 4kg/cm

spring constant k is estimated on the following considerations;

Table 2 Parameters of 3-storied Structure analyzed

t-lass Spring

Element Mean Value

Coefficient

of Variation

Element Mean Value

Coefficient

of Variation

mi

cm

0.130 *''"/sec2 0.17 h

^jy g ton/cm

0.1

m2 0.198 0.17 k2 418.6 0.1

"13 0.204 0.17 k3 363.0 0.1

"o 0.400 0.17 ks 1300 0.2

I 3g373ton.cm.sec2 0.17 kR 2.166xl0^^°"*'='" 0.2

Considering a structure having a relatively rigid horizontal members, spring constants of

each story of the structure, k, are represented by

i =1

where k^ are the values of horizontal rigidity of each vertical member at a corresponding

story and these can be obtained by the method of D-value. Applying the formulas of "the

First Expansion" to Eq. 45, the mean value and variance of the spring constant k are represented by

VI-12
_ N k = J

I = 1

Var(k)= 2 Var(A,)

1-1

Therefore, the coefficient of variation 5 of the spring constant k is represented by

Considering that the mean value and variance of all vertical members have

the same value, Eq. 47 becomes

Ok =VN Var(*) /Hk = o^/VN, (48)

where is the coefficient of variation for the horizontal rigidity of each vertical

member at a story. Based on the statistical arrangement of materials, the coefficient of

variation 5^^ for the horizontal rigidity of reinforced concrete columns is estimated as 0.3.

The values of the coefficients of variation, shown in Table 2, were obtained by application

of Eq. 48 in which 0^^=0.3. The values of the coefficients of variation for sway and rocking spring constants, k and k , are 0.2. These values were estimated from the variation

of the natural period among many standard 5-story apartment houses constructed by the

Japan Housing Corporation. The values of the coefficients of variation for mass were determined by the relationship which is obtained by application of the formulas of "the

First Expansion" to the equation of eigen value on a single-degree-of-freedom system. In

this case, the coefficient of variation for the natural period of the system is assumed to

be 5^=0.1. The mean value and coefficient of variation of the damping constant, for a fundamental mode used in this example, are h =0.06 and d, =0.3, respectively. These were

1 hi

determined by a statistical analysis of the materials concerning damping constants of

many actual structures. Finally, the properties of the earthquake ground motions must be

determined. In this example, it is assumed that the structure is constructed in Muroran,

where many strong earthquakes have been observed. Fig. 3 shows the power spectra of a ground accelerations observed in Muroran Harbor. However, a simulated power spectra

corresponding to those of the observed earthquakes can be obtained by use of Eq. 14, where

parameters co^, E, and s in Eq. 14 are assumed as stochastic variables. Fig. 4 shows the

simulated power spectra. These two figures. Figs. 3 and 4, are quite similar to each other.

Therefore, the simulated power spectra shown in Fig. 4 were utilized in the following

niomerical calculation. The prescribed mean value and variance of parameters co^, E, and s on

the simulated power spectra are shown in Table 3. The standard deviation of the ground

VI-13
Table 3 Constants for Simulated Power Spectral Density Function

of Earthquake Ground Motion at Muroran Harbor

Mean Value Coefficient of

Variation

Predominant Frequency

OJg T5_57rad/sec

0.16

Damping

0.235 0.08

Power Level

S

38.559^1^/-^

......

0.17

motions, having this simulated power spectra, 0.. , is 70 gals. The maximum acceleration is

xg

about 3 times as great as the standard deviation in this case, namely 210 gals. The mean

value and coefficient of variation for a time duration of the ground motions are assiamed to

be T=30 seconds and 5 =0.2.

T

Fig. 5 shows the stochastic distribution of the maximum responses. The solid lines in

the figure show the mean values of the maximum displacements. The broken chain lines show

the confidence band located on either side of the mean value at a distance of the standard

deviation and the maximum displacements corresponding to 95% probability. The probability

distribution of the maximum responses is assumed to be Gumbel's First Asymptotic Distribution. The coefficient of variation of the maximum displacements and the ratios of the mean maximum displacements, corresponding to 95% probability to the maximum displacements,

are shown in Fig. 6. The coefficients of variation of the maximum displacements are about

0.3 to 0.4 and the maximum displacements, corresponding to 95% probability, are about 1.8

times as much as the mean maximum" displacements . This fact shows that the stochastic variation of the earthquake response of a structure is extremely important in designing a structure with proper safety against earthquake ground motions.

Effect of Stochastic Variables to Variation of Earthquake Response of a Structure

The structure model used in the previous section is examined again in this section.

The coefficients of variation of the stochastic variables, used in the previous section,

are employed as the standard values of these coefficients of variation. Varying independently all the coefficients of variation of each one of all stochastic variables, in the

range of 0.05 to 0.4, gives the results shown in Fig.s 7 to 9 . The significant features

in these figures are as follows:

VI-14
(1) The most influential factor in the variation of the maximum displacements is the

variation of the predominant frequency co^ of the earthquake ground motions

.

(2) The variation of the spring constants k of the structure is the most influential

factor, as well as the predominant frequency ca of the earthquake ground motions,

g

to the variation of the maximum story displacement,

(3) The variation of mass m, fundamental damping constant h^ of the structure and the

damping constants C in the power spectra of the earthquake ground motions are not

significant.

(4) The variation of the spring constants of the structure is not influential in the

variation of the maximum response of the foundation. Similarly, the variation of

the sway and rocking constants is not influential in the variation of the

maximum response of the structure, except for the foundation.

(5) The variation of the duration of time of the earthquake ground motion is not an important factor.

In this example, the predominant period of the earthquake ground motions is 0.4 seconds,

while the fundamental period of the structure is 0.38 seconds. Therefore, it is natural

that the variation of the predominant frequency of the earthquake ground motion is an

important influential factor with respect to the variation of the maximum response. Hence,

it must be recognized that the feature described in (1) is not general. It is expected that

the variation of the predominant frequency of such earthquake ground motions as white noise

excitations are not effective with respect to the variation of the maximum response.

Conclusion

A study of the stochastic earthquake response of a structure, which consists of a

nondeterministic stochastic variables, is explained in this paper. If the statistic

properties of the nondeterministic variables relating to the problem of earthquake

responses of structures can be obtained with sufficient accuracy, then the response

properties of a structure subjected to earthquake ground motions can be reasonably evaluated

by the stochastic theory described herein.

VI-15
References

1. Emilio Rosenblueth: "Some Applications of Probability Theory in Aseismic Design",

Proceedings of the 1st World Conference on Earthquake Engineering, 1956.

2. Hiroshi Tajimi: "Earthquake Responses of Structures Estimated by Statistical Method",

Transactions of the Architectural Institute of Japan, No. 603, 1958.

3. M. Ozaki, M. Watabe, M. Hirosawa, Y. Matsushima and Y. Yamazaki : "Earthquake Response

Prediction of Reinforced Concrete Buildings", Transactions of the Architectural

Institute of Japan, No. 171, 1970.

4. T.K. Hasselman and G.C. Hart: "Modal Analysis of Random Structural Systems", Journal

of the Engineering Mechanics Division, Proceedings of the American Society of Civil

Engineers, June, 1972.

5. M. Hoshiya and H.C. Shah: "Free Vibration of Stochastic Beam-Column", Journal of the

Engineering Mechanics Division, Proceedings of the American Society of Civil Engineers,

August, 1971.

6. Yutaka Matsushima: "Perturbation of Earthquake Responses for Systems Composed of

Stochastic Variables", Transactions of the Architectural Institute of Japan, No. 210,

1973.

VI-16
Table 1 Actual Variable Factors Corresponding to Probabilistic Variable rj^

Duration

Time ft

Earthquake

Property

nAmnino ' 1 Mass ro Spring

n

T

1 '"3L+2 hi" 1 Ml ''3L+5-»N kl

=1 '•3L+4 1^2

V* A Si 1

1

»"5 «2

""6 0)92 n S2 —f—

Mn '3L+4+2N

1 B '3L+3-»N Mo r3L+5+2N

rSL-l rSL+a+N I '•3L>6»2N

'•3L rSL+l

VI-17
VI-18
VI-19
a) for Max.

Displacement

b) for Max. Story

Displacement

c) for Max. Story

Displacement subtracting

Rocking Effect

r /

I

_§

3F i.

2F

IF

3F

2F

IF

I 1.

Fig.6 Coefficients of Variation «|Y|mix and Ratio of Maximum Response

corresponding to Probability 95%to Mean Maximum Response

(|YU)0.95/ |Y|

Fig.7 Relation between Coefficient of Variation of Maximum Displacement and of Variables

on Input and System

VI- 2 0
er

2

ou

u ^

o X

(a)

- BF

.2

(e)

.1

.3 .4 <^

( b)

BF

:3F "2F

IF

.4

.2

(f )

.3 i

.3 .4

(c)

BF .6

.3 .4

(9)

er

(d)

.2 .3 .4"?

1^.

Coefficient of Variation using as Standard Value

Fig.8 Relation between Coefficient of Variation of Maximum Story Displacement and of

Variables on Input and System

w c

(0 Of

o !d S

1

1"

«

U ^

Z u

U If)

.2

.4 (7„

.8

.4 3F

2F,1F

.3 .4 a.

.3 .4 ffr .» .2 .4 0-„,

3F

2 F,1F

.2

.1 .2 .3 .4

.1 .2 .3

.1 .2 .3 .4 ff„,

Coefficient of Variation using as Standard Value

Fig.9 Relation between Coefficient of Variation of Maximum Story Displacement

subtracting Rocking Effect and of Variables on Input and System

VI-21
LEAST WEIGHT STRUCTURES FOR

THRESHOLD FREQUENCIES

by

R. D. McConnell

Graduate Student

Catholic University

Washington, D.C.

ABSTRACT

This paper demonstrates a method for designing least-weight, or optimized, structures

when a threshold frequency (within appropriate force factors) is known. The process is

highly amenable to a job-by-job application due to its simplicity. Small mathematical

models of a structure can be analyzed by a hand-calculator while standard static-load

finite element programs can be used for large structures.

Key Words: Least weight, optimization, structures, frequency response.

VI-22
Introduction

Aerospace designers have long been designing component structures to a minimal-fundamental, or "threshold", frequency specification. In general, dynamic force levels imposed

in such designs are determined by the principal inertial force excitations to which the

structure under design will be sxibjected at frequencies known to be below the specified

"threshold design frequency". Undoubtedly such concepts may have been used in various

earthquake-related analyses and designs . More formalized procedures as specified by various codes, however, fall into two categories: a table of component forces (which often

result in grossly conservative or grossly underdesigned structures) ; or sophisticated

computer analyses using detailed models which undergo dynamic time-history force simulations.

Such sophistication is well known to be expensive, time-consuming and somewhat arbitrary in

earthquake analyses due to the approximations used as the forcing functions. Such large

programs often have the hazards of inadequacies in the mathematical algorithms and error- conditioning problems with large models.

Although it may not appear as generally applicable for ground-supported structures, the

design of svibstructures for earthquake induced forces could use such aerospace methods.

By "s\±istructures" here is meant any component structure for which the predominant frequency to which it may be subjected is known to a reasonable degree of accuracy. For example, radio towers, watertanks, electrical gear, etc. which are supported by larger

structures will be subjected to the natural frequencies of the "base structure" regardless

of the earthquake spectrum. (Technically, "larger" structure would imply considerably

higher dynamic modal-mass) By knowing the approximate amplification factors for the

principal contributing frequencies (and corresponding base-excitation levels) of the

"base structure," a substructure specification for the 1) dynamic amplification factor,

2) arbitrary safety-factor and 3) threshold design frequency can be determined. (This has

assumed consideration, where necessary, for an assumed mass-spring representing the substructure as part of the "base structure" model.) The arbitrary factor-of-safety would

take into consideration the spread between the specified threshold design frequency and

the fundamental (or major contributing) frequency of the "base structure", in addition to

the assumed possible accrued errors in the particular analyses

.

For these reasons, this paper demonstrates a method for designing least-weight, or

optimized, structures when a threshold frequency (with appropriate force factors) is

known. The process is highly amenable to a job-by- job application due to its simplicity.

Small mathematical models of a sturcture can be analyzed by a hand-calculator while

static- load finite-element programs can be used for larger structures. Design (memberselection) methods, whether by hand or by special program algorithms, becomes the same

common set of procedures as would be introduced in any optimum-design methods.

Structural Optimization Methods

With the increasing availability and intricity of finite-element matrix programs for

structural analysis, there has been a growing interest in, and development of structural

optimization procedures to augment those programs. They generally optimize by minimizing

weight and/or cost or a similar objective function. Finite-element programs use matrix

VI-23
representations, or idealizations, of real structures for the purpose of behavioral analyses

under various loading conditions and constraints.

Most methods and applications involving optimization of dynamic properties and behavior

of structures have, in general, incorporated the basic procedures of prior studies in optimization under static load conditions.

The goal in structural optimization is to minimize a prescribed merit function. This

merit function could be weight, cost, fabrication criteria, or a combination of these (7)

.

Active constraints on the minimization of the merit function (or "optimality criterion")

would be the analysis equations for the set of design variables, such as stress solutions

for the design of the members. Side-constraints would be limits on displacements, minimim

member sizes, etc. These constraints must not be violated when optimizing the mathematical

representation of the structure to a particular merit-function. In those cases where the

geometry is to be optimized, volume or space constraints may be imposed.

Structural optimization methods can be divided into two general categories : mathematical programming involving search techniques , and optimality procedures . The principally

identified methods in the first category are feasible-usable direction (16) , gradient

projection (17) , optimum vector (8,9) , linear programming (15) and allocation (13) , all

being defined as explicit minimization processes . Implicit minimization methods in the

same general mathematical search category are steepest descents (18) , adaptive gradients

(19) , variable metric (20) and conjugate gradients (21) . These identities are indicative of

the redesign search routine algorithm. "Implicit" methods use penalty functions (14) to

transform a constrained minimization problem to an unconstrained one. Conceptually, these

can also transform discrete values of the design variables into general unconstrained

functions

.

The optimality procedures include fully-stressed deisgn (22, 23, 24, 25, 26, 27, 4), uniform strain-energy density (28, 29, 27), simultaneous buckling (30, 6), and limit design

(10, 12, 3) . Most examples demonstrated in the mathematical search technique group have been small,

simple models of structures, generally involving static loading conditions. These procedures are very time-consxmiing, may not monotonically converge and involve much programming

effort (8, 9, 16)

.

Virtually all practical applications involving larger model representations of real

structures have employed one of the optimality methods. Conceptually, the optimality

methods require solution of first derivatives of the merit function, or a substitute function

in which case the method is classified as a sxibstitute optimality process. (For example,

the use of fully-stressed convergence is substituted for minimxmi weight.) In the actual

iterative procedures, derivatives are not calculated specifically, hence the redesign calculations are relatively few. In the mathematical search technique, however, second derivatives must be extracted or approximated at each iteration to determine sensitivity coefficients for redesign toward a minimum of the merit function.

VI-24
The two siibstitute optimality methods which have been used with success are fullystressed design and uniform strain-energy density design. These have been used for linearelastic design structures. With the exception of such techniques as penalty functions (14)

in the mathematical search procedures, in general, with the exception of specific sideconstraints, must have equivalent linearized functional representation. Such linear representation of discrete values are generally approximate and cannot always be so expressed.

In the realm of optimization of structures, the penalty function, or unconstrained

optimization, techniques have not been very efficient. Such penalty functions introduce

near-singularities in the search process with consequent instabilities (7)

.

Most literature on applications and comparative reviews conclude that the more efficient and successful methods have been those of the optimality, or substitute optimality

category.

It has long been known that simple structures optimized under a one-load condition will

converge to a determinate structure in the absence of side-constraints which maintain mini- mum sizes for all members in the original structure (25) . (One frequency constraint is

considered the dynamic equivalent of a one load-condition.) That determinate structure

will have member sizes determined only by the forces in the members. The allowable stress

could be prescribed by strength, yield or buckling criteria, and may be different for each

member.

The two methods used today for practical size problems under dynamic constraints or loading conditions are, as in the case of static loads, the siobstitute optimality approaches

of fully-stressed design and uniform strain-energy density design. These two produce the

same optimum structure when the structure is made of only one material.

In the case of resonant frequency optimization by fully-stressed methods, the allowable

stress is uniform for all members in the structure. This is to say that if strength

criteria do not preclude the adjusted stress-level, fully-stressed methods in the dynamic

resonance case will reduce the stresses uniformly (thereby increasing sizes) ,to provide the

stiffness sought. This step is often called "proportioning". Re-analysis is then necessary

as the mode-shape, and consequently the effective force-distribution, changes. As noted

earlier, side-constraints may be introduced to preserve strength in particular members.

The primary use of structural optimization procedures in dynamics problems has been

to optimally adjust the fundamental frequency of the structure while assuring all additional

strength requirements.

One reference (25) , is of particular note here in that it is a representative practical

application of optimizing a real structure to a fundamental frequency threshold level by the

fully-stressed substitute optimality approach. The authors. Young and Christiansen, point

out that they do not find an absolute, or global, optimum, but they do produce a "good,

efficient structure."

Their procedure was essentially: 1) produce an original design; 2) analyze for

resonant frequencies in the reduced models , mode shapes and corresponding modal forces in

the detailed model; and 3) redesign (or proportioning). These three steps, basic design,

analysis and redesign, are the fundamental steps common to all iterative optimization mettiods.

VI-25
This paper describes a simple, stable method for adjustment of the entire frequency set

to a minimum level. The structure adjusted by this efficient method will be simultaneously

optimized to a least weight criterion without requiring the use of eigenvalue/eigenvector

extraction routines, or algorithms to transpose mode shapes to forces. This procedure con- verges rapidly and requires only standard finite-element programs for static-load conditions.

Procedure

For those cases where the entire set of eigenvalues is shifted to provide a fundamental

frequency above a prescribed minimum level, proportioning of the structure is possible.

Methods for total adjustment with proportioning have been presented in the literature as described earlier. It is often possible, however, to obtain such results by a rapidly

converging, stable process using finite-element programs written for static- load analyses.

The economy so realized, as contrasted to the use of eignevalue extraction routines, permits

the use of larger, more detailed models in these instances. The element of instability of

comparable methods in the literature is principally inherent in the algorithms used to

transpose mode-shapes to relative forces in the structural elements (25, 4). Modal extraction and transposition are not explicity required in the familiar Rayleigh energy method to

obtain the fundamental frequency of a structure. Also, by using the more detailed model

of a structure, rotary inertias generally become insignificant. Their exclusion in a Rayleigh approximation, consequently, does not seriously affect the results.

Static equivalent loads which can be factored by any arbitrary constant, are imposed

by ratio and direction in accordance with the assumed (approximate) fundamental mode-shape.

Resulting forces are then extracted by the routine stiffness matrix procedures for static

loads. The force set is then linearly proportioned to achieve a fully-strssed state in

that member which experiences the numerically largest force using the prevailing size of

that member. All members are then simultaneously proportioned to the forces and scaled

by the ratio of A /A where X is the desired eigenvalue level and X is the currently exr E F E

tracted approximation. The process is repeated using applied forces proportional to the

latest mode- shape. The iteration is complete in the Rayleigh method when the X extracted

E

approximates X^ desired.

This modified Rayleigh method is simultaneously converging to the true-mode shape of

the fundamental mode while being proportioned accordingly.

In most cases, an approximate mode-shape is known. When this is not the case, it would

be advisable to use a procedure such as the "inverse power method with shifts" to extract

the fundamental frequency and its corresponding mode-shape for only the first step^'''"'"^

Following adjustments, the routine Rayleigh method, as modified above, may then be used.

Convergence can be assured by checking the total structural weigh t upon each successive

redesign step. If there is a divergence due to a poor estimated mode-shape, it is only

necessary to carry the Rayleigh iteration to successive steps (without proportioning) until

convergence is assured. The advantages of this method, in contrast to those in the

literature, are 1. since only the fundamental frequency is of concern, the Rayleigh method

is efficient and economical; and 2. the element forces are already known during each step

to the process.

VI-26
The following is a statement of the basic Rayleigh iterative steps.

0)2 =

Z F . (t)" ri r r=l

A" E M (d) )

, r r=l

Where F^^ are the inertia forces corresponding to an assumed (or prior calculated)

mode shape

.

A" ())" are the deflections due to F . where 6" is the normalized vector; (5) . r ri

Example

A simple 3x3 matrix model is shown in Figure 1. The schematic spring model corresponding

to the structural model is identical to one used in Reference (5) . The first iteration

cycle is identical. The s\±>sequent steps toward weight/frequency optimization can be com- pared with the elementary method, in that reference, whereby only the extraction of the

original fundamental frequency was required.

M,=2 Ib-sec^/in —*- (772 lb)

2 M-=l lb-sec /in

C386 lb)

2 M3=l lb-sec /in ^ (386 lb)

TTTTTTr

5

ti •H

o o O o o o o o VO CM

nIC

jVSA/I "^1

Figure 1

3 Degree-of-Freedom Example of Optimal Adjustment

of Fundamental Frequency by Rayleigh Method

Table 1 illustrates the steps in a simultaneous Rayleigh extraction of the fundamental

frequency of this structure while proportioning members to optimized its weight. It con- currently shifts the fundamental frequency to a prescribed threshold level of O) = 30 rad/sec.

The assumed mode shape is obtained by a Ig. force on each mass. The first cycle produces a frequency of 28.2 rad/sec, and a mode shape of 1, 2.02 and 3.22. The spring stiffnesses (used in the Fig. 1 schematic to illustrate the method) are then adjusted to 6000,

4340, and 2670 lb/in to result in a uniform stess. This is obtained in this case by assuming

the effective spring lengths equal and spring stiffnesses proportional to area. If a spring,

therefore, has twice the area, it has twice the weight (for the same lengths) and will take

twice the force at the same stress. The mode shape to which this model must coverge can be

recognized to be 1, 2, 3 for these conditions. That "computed mode shape" is achieved in

VI-27
second cycle following proportioning of the springs. That cycle produced a frequency

0) = 29.5 (increased due to the basis of the spring proportioning).

The third cycle is exact, in this instance, as the assumed and computed mode-shapes are identical. The resulting frequency is 29.3. An adjustment by oji/f^i of the stiffnesses to

produce 30. rad/sec results in the minimum total weight springs as shown.

Conclusion

In conclusion, the above example has demonstrated "fully-stressed" optimization con- current with threshold-frequency adjustment by a procedure so simple as to permit the

solution, in this instance, by the use of a hand calculator. The use of a model used

earlier in texts (to demonstrate the Rayleigh method) was to permit comparison with this

paper's augmented method. In addition, the particular model was thought to be of an

"optimal" variation in member sizes (for first-mode stresses) at the beginning of the process. Further efficiency in weight was obtained.

The possible applications to earthquake design of sub-structures become most apparent

upon reviewing the reasons for such requirements and applications in the design of aerospace structures

.

VI-28
References

1. McConnell, R. D. : Direct Optimization of Dynamic Response of Structures by Spectral

Synthesis. Ph.D Thesis, Catholic University, 1974,

2. Gere, J. M. ; Weaver, W. : Analysis of Framed Structures. Van Nostrand, Princeton,

New Jersey, 1965.

3. Hodge, P.; Prager, W. : A Variational Principle for Plastic Materials with Strain

Hardening. Jour. Math. Physics, Vol. XXVII, 1948, pp. 1-10.

4. Rubin, C. P. : Minimum Weight Design of Complex Structures Subject to A Frequency

Constraint. AIAA Journal, Vol. 8, No. 5, 1970.

5. Biggs, J. M: Introduction to Structural Dynamics. McGraw-Hill Book Co., New York, 1964.

6. Shanley, F. R. : Weight-Strength Analysis of Aircraft Structures. McGraw-Hill Book Co.,

New York, 1952, pp. 1-260.

7. Melosh, R. J. : A Plan for Computer-Assisted Optimization of Structiires. NAS 5-11779,

NASA, G.S.F.C., Greenbelt, Md. 1970.

8. Gelatly, R. A.: Development of Procedures for Large-Scale Automated Minimum-Weight

Structural Design. AFFDL-TR-66-180 , Air Force Flight Dynamics Laboratory, WrightPatterson AFB, Dec. 1966.

9. Gelatly, R. A.; Burno, J. A.: Development of Procedures for Large-Scale Automated

Minimum-Weight Structural Design, Supplement 1, Programmer Documentation. AFFDL-TR-66-

180, Supplement 1, Air Force Flight Dynamics Laboratory, Wright-Patterson AFB,

Dec. 1966.

10. von Mises, R. : Mechanik der festen Korper im Plastish-deformablen Zustand. Nachr.

Kgl. Ges. Wiss. Gottingen, Math-Physic Klasse, 1912, pp. 582-592.

11. McCormick, C. W. , Editor: The NASTRAN User's Manual. NASA SP-221, National Aeronautics

and Space Administration, Washington, D.C. Oct. 1969.

12. Nadai, A.: Plasticity, A Mechanics of Plastic State of Matter. McGraw-Hill Book Co.,

New York, 1931.

13. Melosh, R. J.; Luke, R. : Approximate Multiple Configurtion Analysis and Allocation for

Least Weight Structural Design. AFFDL-TR-67-59 , Air Force Flight Dynamics Lab.,

Wright-Patterson AFB, 1967.

14. Fiacco, A. V.; McCormick, G. P.: Computational Algorithm for the Sequential Unconstrained Minimization Technique for Nonlinear Programming, Part II. Management

Science, Vol. X, No. 4, pp. 601-617.

15. Pope, G. C: Application on Linear Programming Techniques in the Design of Optimum

Structures. AGARD Symposium on Structural Optimization, Istanbul, Oct. 1969.

16. Zoutendijk, G. : Method of Feasible Directions. Elsevier Pi±>lishing Co., Amsterdam,

1960.

17. Rosen, J. B.: The Gradient Projection Method for Nonlinear Programming. Jour. SIAM

Appl. Math., Vol. IX, No. 4, Dec. 1961, pp. 514-532.

18. Cauchy, A. L.: Methode Generale pour la Resolution des Systems d' equations simultanee.

Compes rendu de I'Academie des Sciences de Paris, Vol. XXV, pp. 536-538, 1947.

19. Rosenbrock. H. H.: An Automated Method for Finding the Greatest or Least Value of a

F\anction. Computer Jour., Vol. Ill, 1950.

VI-29
20. Fletcher, R. ; Powell, M. J. S.: A Rapidly Convergent Descent Method for Minimization.

Computer Jour., Vol VI, 1963, pp. 163-168.

21. Fletcher, R. ; Reeves, CM.: Function Minimization by Conjugate Gradients. Computer

Jour., Vol. VII, 1964, pp. 149-154.

22. Maxwell, J. C: Scientific Papers II. Cambridge University Press, 1890, pp. 175-177.

(Reprinted by Dover Publications, New York, 1952.)

23. Michell, A. G. M. : The Limits of Economy of Material in Frame-Structures. Phil. Mag.

(London), Ser. 6, Vol. VIII, No. 47, Nov. 1904, pp. 589-597.

24. Cilley, F. H. : The Exact Design of Statically Indeterminate Frameworks: An Exposition

of its Possibility, but Futility. Trans. ASCE, Vol. XLIII, June 1900, pp. 353-407.

25. Young, J. W. ; Christiansen, J. N. : Synthesis of a Space Truss Based on Dynamic

Criteria. Jour. Struct. Div. ASCE, Vol. SCII, No. ST6, Dec. 1966, pp. 425-442.

26. Gellatly, R. A.; Gallagher, R. H.; and Luberacki, W. A.: Development of a Procedure

for Automated Synthesis of Minimum Weight Structures. FDL-TR-64-141 , WrightPatterson AFB, Ohio, 1964.

27. Venkayya, V. B.; Khot, N. S.; and Reddy, V. S.: Energy Distribution in a Optimum

Structural Design. AFFDL-TR-68-156, Wright-Patterson AFB, Ohio March 1969.

28. Prager, W.; Taylor, J. E.: Problems of Optimal Structural Design. Trans. Jour.

Appl. Mech. ASME, Vol. XXXV, Series E, No. 3, Mar. 1968, pp. 102-106.

29. Taylor, J. E.: Maximum Strength Elastic Structural Design. Jour. Eng. Mech. Div.

ASCE, EM3, June 1969, pp. 653,663.

30. Gerard, G. : Minimum Weight Analysis of Compressive Structxares. New York University

Press, New York, 1960.

VI-30
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VI-31
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VI-32
DUCTILE SHEAR WALLS IN EARTHQUAKE-RESISTANT MULTISTORY BUILDINGS

by

Mark Fintel

Director, Engineering Services Department

Portland Cement Association

Skokie, Illinois 60076

ABSTRACT

Slender shear walls in multistory buildings are discussed in this paper and answers are

given to such questions as "Why do we need them?" and "What do we know about their design?".

Also discussed are the historical development of the use of shear walls, their performance

in earthquakes of the past 10 years (both good and bad) , and finally, the available design

information and our future needs in the area of design of shear walls for strength, stiffness

and ductility, as well as needs in the area of analysis for the dynamic response of shear

wall structures.

Key Words: Buildings; Damage; Dynamic Response; Earthquakes; Safety; Shear Wall; Stiffness.

VI-33
slender shear walls can be defined as vertical cantilevers, with various cross sections

such as: rectangular (a vertical plate), I, box, and other elevator walls. The shear walls

support the vertical load, in addition to their function to stiffen the frames in their re- sistance to lateral loads due to winds, earthquake or blast. Although interior and exterior

concrete walls have been used to stiffen structures as long as reinforced concrete itself

has been in use, the modern concept of shear walls designed as vertical slender cantilevers were first utilized in 1948 in housing projects in New York City and in Chicago in buildings

designed for wind forces, to augment the lateral resistance of the frames.

In nonearthquake areas of the United States and Canada concrete buildings with more

than 15 to 20 stories are usually designed with shear walls, primarily to improve their

stiffness. Economically, the inclusion of shear walls seems to be the least expensive way

to increase the overall rigidity of concrete buildings, since they serve the triple function:

to support gravity loads, to provide lateral resistance, and to function as a wall.

Analysis for lateral loads of buildings containing shear walls was carried out initially,

in the 1950s, by assigning all the lateral loads to the shear walls, since it was felt that

the very big difference in stiffness between the shear walls and the frame would cause the

shear walls to accept the total lateral loads. This inaccurate assximption may have been

conservative for the computation of shear wall moments; it is, however, not conservative for

the frame, and particularly in the upper parts of the building.

Formal procedures for shear wall-frame interaction were first introduced in the early

sixties. Fig. 1 shows the concept of the interaction, with the resulting internal forces,

which substantially increase the overall stiffness of the combined system.

Most of the recent prominent ultra-high-rise reinforced concrete buildings were built

without any additional cost for the lateral resistance, which was mostly accommodated within

the 33 percent increase in the allowable stresses when lateral loads are considered. The

high lateral rigidity was achieved as a result of the shear wall-frame interaction.

Performance in Earthquakes

To judge the merits of shear walls for earthquake resistance, an examination of their

performance in earthquakes should be made, and particularly the comparative behavior of

frame buildings and buildings containing shear walls during the earthquakes of the last 10

years, starting with the earthquake of Managua, and going back in chronological order to the

other earthquakes.

Managua Earthquake, 1972

The earthquake of Managua of Dec. 23, 1972 seems to be the most significant of the re- cent earthquakes,"*" since side by side were two buildings (Fig. 2) representing the two

different structural systems. The building on the left is the 15-story Banco Central which

is principally a frame building, while the building on the right is the Banco De America, an 18-story shear wallframe interactive system.

The Banco Central was designed in the early 1960s; the structural system of the tower

(Fig. 3) is a one-bay frame with two small reinforced concrete cores and an infilled wall at

the end of the building. These stiff elements may have done the structure more harm than

VI-34
good by introducing torsion due to their off-center location.

The building was subjected to violent shaking, as could be observed from the damage of

the nonstructural elements, both inside and outside (Fig. 4). Except for the 4th floor

steel roof over the auditorium which fell off its support, the only structural damage observed in the building was some ripping between the off-center core and the slab. The structural frame of the tower itself showed little distress. There was evidence of yielding at

some junctions between the columns and the beams; however, the structure had enough ductility

(whether designed for it or not) to sustain the large distortions. Inside, however, the

building was in shambles over most of the stories, due to the extremely violent shaking ex- perienced by the building.

In contrast, the building across the corner, the 18-story Banco De America, (Fig. 2)

exhibited an entirely different perfmance, both inside and outside. The plan of the building contains four centrally located cores, arranged symmetrically within the peripheral

column system as shown in Fig. 5. The cores are interconnected by two rows of stocky beams,

many of which are penetrated by ducts. A general comment: the number of shear walls in

this building is substantially higher than is usually provided in wind-resisting buildings

of comparable height and plan size.

The interconnecting beams between the cores suffered repairable shear damage through

most of the height of the building. Observing the large amount of flexural reinforcement

in the beams, it should be recognized that it is almost impossible to provide enough shear

capacity (whether the beams are penetrated by ducts or not) to be able to develop flexural

hinging at the ends of the beams.

There was very little evidence throughout the height of the building of any violent

shaking. All the furniture was in place, and there was no discernible nonstructural

damage

.

A comparison of the above two buildings shows that both were well designed and well

constructed, within several years of each other. Both buildings were designed according to

the United States west coast standards in force at the time of their design. Although both

buildings were subjected to the same earthquake motion, one had very severe architectural

damage; the other could be reoccupied immediately while the repairs proceeded. There was

only one difference between the two buildings - a healthy system of shear walls which restricted the interstory distortions, thus providing damage control.

Figs. 6 and 7 show another pair of comparative buildings from the Managua earthquake.

Both of these 5-story buildings were located in areas of high damage. The 5-story Insurance

Building (Fig. 6) outwardly appeared so badly damaged that people did not trust themselves

to enter it for several days. Later inspection showed that, although the exterior masonry

walls and the interior partitions were badly damaged, there was little structural damage to

the building's moment-resistant frame.

In contrast, the 5-story Enaluf Building (Fig. 7) which has a relatively large reinforced concrete core in addition ot the frame, went through the earthquake exceptionally

well. The only structural distress was slight horizontal cracking in several of the exterior

first story columns, and a shear wall on the ground floor had some damage adjacent to a duct

VI-35
penetration. Otherwise, neither structural nor serious nonstructural distress appeared in

the building.

Performance of Other Modern Frame Buildings in Managua

There were also a number of other well-designed moment-resisting frame buildings in the

6-8 story range. They were all subjected to intense shaking and large distortions as evidenced by the severe damage to their nonstructural components and finishes. All of them

exhibited sufficient ductility, since there was almost no structural distress. They all

performed according to the present code philosophy - little or no structural distress, however, quite a bit of nonstructural damage.

The 8-story Supreme Justice Building had minimal structural distress, but the inside was in shambles.

The 8-story Social Security Building (Fig. 8) outwardly showed no distress except for

the collapsed roof over the elevator machine room, and little structural distress on the

inside. However, the nonstructural damage was considerable and the staircases were full of

partition debris and could not have been used to evacuate people had the earthquake occurred

during the daytime.

The 8-story Telecommunications Building had post-tensioned beams on the column lines

spanning across the entire building. There was on off-center core at the far end of the

building was in evidence; however, the post-tensioned beams of the tower showed (with one

exception) no substantial structural damage.

Summarizing the behavior of this group of frame buildings , it is evident that the

buildings had enough ductility, whether intentionally designed for it or not, to sustain the

large distortions to which they were subjected during the earthquake. The high degree of

economic damage these buildings suffered was apparently not so much from the ground shaking

as from an inadequate design philosophy - a philosophy in which we design the structure for

large distortions, but we do not detail the rest of the building (which in many cases com- prises up to 80 percent of the value) to accommodate the large distortions without damage.

Of particular interest was the performance of the National Theatre, built several years

ago in the style of the Lincoln Center in New York. The structure contains a U-shaped reinforced concrete shear wall around the auditorium within another U-shaped shear wall of

columns and beams infilled with 18-in. thick solid masonry around the lobby and stage. Except for a few marble statues which were thrown off their pedestals, there was no evidence

that the building had just gone through an earthquake. Neither structural nor nonstructural

damage could be found.

San Fernando Earthquake, 1971

The Indian Hill Medical Center is an example of an acceptable earthquake performance of

a shear wall-frame type building. The building was restored and put back into operation

within a short period after the earthquake. The structure of the Indian Hill Medical Center

consists of beam column frames supplemented by shear walls. The shear walls exhibited some diagonal cracking and other local distress; they were repaired by increasing their thickness.

On the compound of the Veteran ' s Administration Hospital , where some buildings of 1920

vintage collapsed, several auxiliary buildings built as reinforced concrete boxes went

VI-36
through the earthquake without structural damage; even the chimney of the central boiler

plant had only some sliding at a cold joint. Unfortunately, these cases of excellent behavior somehow escaped the attention of the profession, which was busy examining the collapses . Caracas, Venezuela Earthquake, 1967

Fig. 9 shows a multistory building with a very flexible skeleton of the type prevalent

2 in Caracas. There were no shear walls used in the Caracas buildings. The skeletons were

filled with brittle and weak hollow clay tile infill walls. During the earthquake the

buildings were subjected to large distortions and the weak partitions exploded, as shown in

Fig. 10. It is obvious that such damage should have been expected from buildings with

flexible skeletons undergoing large distortions and filled with brittle partitions.

The 17-story Plaza One Building was the only complete shear wall building in Caracas.

It was located within an area of extremely high damage. As seen in Fig. 11, one of its

neighbors, a 10-story building, collapsed while the other surrounding structures suffered

severe damage. The Plaza One Building went through the earthquake without any damage whatsoever. The building has shear walls in both directions.

Skopje, Yugoslavia Earthquake, 1963

Many of the residential buildings in Skopje up to 10 stories high had two shear walls,

all across the width, flanking the central stairway. The shear walls were usually non-rein

forced of poor quality concrete. Nevertheless, the rigidity of the shear walls did not per

mit interstory distortions, and consequently, there was no damage. In some instances slip

of the cold joints between successive story lifts was observed.

Fig. 12 shows the 14-story Party Headquarters, which was the only building in Skopje

with a structure similar to our shear wall-frame systems. Being the party headquarters, it was designed more carefully and constructed with greater attention to quality. Three non- reinforced shear walls in the center were of good quality concrete. Although it is known

that the building iinderwent severe shaking during the earthquake (according to witnesses

who were thrown from one end of the room to the other) , there was neither structural nor nonstructural damage in the building, with the exception of the elevators which did not

function after the earthquake.

Misbehavior of Shear Walls

The preceding description of performance in past earthquakes showed many examples of

good behavior of buildings containing shear walls. On the opposite side of the spectrum

there are two distinct categories of misbehavior of shear wall buildings during earthquakes

(a) interrupted shear walls, and (b) brittle linkage between coupled shear walls.

The Olive View Hospital (San Fernando) shown in Fig. 13 is an example of interrupted

shear walls. The upper four stories contained shear walls; however, they were undesirable

in the ground floor due to the architectural layout, and were, therefore, omitted. The

building distorted during the earthquake by more than 2 feet within the ground story. In

retrospect, it seems that in this building the ductility available in the upper stories

was of little benefit because the presence of shear walls did not allow any distortions;

while the extremely high amount of ductility available in the columns of the ground story

I VI-37
did not really do very much good in keeping the building intact, except possibly preventing

total collapse of the ground story.

Discussion

If we look back and review what we have learned from the previous earthquakes, I believe that we were extremely eager during the last decade to verify the viability of the

concept of the ductile moment-resistant frame. Consequently, after each of the past earthquakes, we have introduced some improvement to the ductile moment-resistant frame, i.e.,

after Caracas we increased substantially the overturning moment; after other earthquakes,

other details were modified. However, I believe that we have not thoroughly examined the

basic concept of our earthquake design philosophy as related to reinforced concrete. For,

I believe if we were to make a thorough examination, we would probably find that the ductile moment-resisting frame without shear walls is a relatively poor structural system for

residential and office buildings which contain a lot of nonstructural elements that are not

designed and detailed to accommodate the large earthquake distortions of thQ moment-resistant frame.

A brief look at the history shows that the ductile moment-resistant frame evolved in

the 1950s out of themoment-resistant frame which, at that time, was the only system for

multi-story buildings for both steel and concrete. By adding ductility to the then available system, we created a convenient solution to the problem of earthquake resistance.

However, in the meantime, better and more efficient structural systems for multistory

structures (both in steel and concrete) were developed for widn resistance. Actually, the

last moment-resistant frame utilized in the east for a very tall building was the 60-story

steel framed Chase Manhattan Bank in New York, built in the early 1960s. At that time,

this 50-story building utilized about 45 lb of steel per square foot of floor area (220

2 kg/m ) . Today, 60-story buildings are built with about 20 lb of steel per square foot

2 (98 kg/m ) , while buildings in the 100-story range are built with slightly over 30 lb of

2 structural steel per square foot of floor area (146 kg/m )

.

The emphasis on ductility as the key to survival of moment-resisting open frames led

to the adoption of concrete structures reflecting characteristics more common to steel

buildings, rather than taking advantage of the strength, stiffness, and ductility inherent

in the natural forms to which concrete lends itself, such as shear walls. Recent experience has demonstrated that the twin requirements of safety and damage control can be better

met by structures possessing adequate stiffness, such as shear walls can most economically

provide, when coupled with sufficient ductility or energy-absorption capacity. This is

especially desirable for apartment and office buildings, where considerable nonstructural

damage can result from excessive interstory displacements during an earthquake. When

sufficient lateral stiffness is built into a structure by the introduction of ductile | shear walls, so that large lateral displacements are prevented, it is doubtful if the j connected frame in a frame-shear wall building will ever undergo the distortions which

would call for the ductility which we now design into them. 1

It seems that for uses like parking garages, stadiums, bridges, and the like, the I ductile moment-resistant frame without shear walls is an excellent earthquake resistant I

VI-38
system. However, for apartment and office buildings in which 80 percent of the value is

nonstructural, we should have more damage control than the ductile moment-resistant frame

can provide. ^

Shear Walls As Elements

Shear walls can be classified as (a) short shear walls (h/d, less than about 1/2) , and

(b) slender shear walls (h/d, more than 2) . Short shear walls are mostly governed by their

shear strength, while slender shear walls are cantilever beams controlled by flexure. If

special details are utilized in slender walls, they can be detailed to have sufficient

ductility.

Another possible classification of shear walls is according to the geometry of the

section: rectangular sections, and I-sections representing both I and box sections as used in cores. An important difference between plain rectangular sections and flanged

sections is the degree to which shear (i.e., diagonal tension) contributes to the total

distortion.

The current state-of-the-art shear wall design consists of individual developments on

a number of aspects-only bits and pieces-with a rational approach slowly emerging.

Testing of Shear Walls

Tests of infilled one- and two-story concrete frames were conducted in the fifties.

After a period of nearly 20 years, testing of shear walls was initiated several years ago,

and some of the new studies have already been reported in the technical journals. There

have been a number of experimental investigations of short shear walls to determine their

3 4 5 shear strength at the PCA laboratories with both monotonic and cyclic loading. ' '

Results of some of these studies have formed the background for the shear provisions for

shear walls in the 1971 ACI Code.^ Also, a test series of long shear walls with monotonic

7 loading has been carried out at PCA. Investigations on coupled shear walls were carried

Q out very successfully in New Zealand. A series of dynamic tests on shear wall assemblies

on the shaking table has been intitated and is underway at the University of Illinois. In

addition, several universities are preparing test series on various aspects of slender shear

walls

.

Analytical Investigation of Shear Wall Sections

A series of analytical studies to investigate the strength, stiffness, and ductility

9 of shear wall sections was carried out recently at the PCA. The mathematical model used

for the development of the interaction diagrams of shear wall sections is based on the nonlinear beam theory. The mathematical model for the "computer test series" has the obvious

shortcoming of simulating only monotonic loading, since no model of possible effects on

concrete due to cyclic loading has yet been developed.

Despite this shortcoming, this series of computer studies gives us a better understanding of the influence of the variables affecting strength, stiffness, and ductility of shear

walls. In addition, the mathematical model has the capability to consider confinement in

any part of the section.

VI-39
The nonlinear beam theory assumes the strains to vary linearly across the section,

while the stresses in both the concrete and the reinforcement vary non-linearly according

to their actual stress-strain relationships, as shown in Fig. 14.

Two principal sections were investigated: a rectangular section with uniformly distributed reinforcement and reinforcement bunched at the ends; and an I-section with concen- trations of reinforcement in the flanges. The latter section represents box or I-sections

as used in elevator cores.

Since we are interested in the properties at maximum capacity, rather than at the code

"ultimate" (which is at the first yielding) it is important to model the properties of the

materials, particularly after yielding, as accurately as possible. For the steel reinforce'

ment there is good stress-strain information available; for the stress-strain characteristic

of concrete a complex equation was developed based on all tests reported in the literature.

The complexity of the expression is not objectionable, since it is intended for computer

application; accurate presentation of both the ascending and descending branches of the

stress-strain curve was the most important consideration.

The load-moment interaction diagram shown in Fig. 15 is the envelope of maximum

capacity, at whatever strain this may occur, rather than for the customary concrete strain

of 0.003. It is found that the strain of the concrete at which maximum capacity occurs increases as the eccentricity increases, from about 0.002 for the compression controlled

branch of the interaction diagram to three to four times as much for the tension controlled

branch (zero axial load). The reason for this increase: as the steel stretches in yielding,

the concrete strain increases progressing along the descending branch; the strain hardening

of the steel causes an increase of the strength of the section. Computation of design

strength based on concrete strain of 0.003 for all eccentricities is conservative. The

deformation capacity of the section from the onset of yielding of reinforcement until its

final rapture, or until the compression failure of the concrete, is the section ductility.

Also, the available section ductility was investigated as related to the axial load

level. The plot in Fig. 16 of sectional ductility against axial loads shows that the Isection with confined flanges has a substantial ductility even at balanced load level.

Obviously, all these analytical findings will have to be confirmed by laboratory tests.

Needed Studies ^ Further work is needed in both the analytical and experimental areas to supplement the

present knowledge for a complete procedure for earthquake resistant shear wall type buildings. Extensive shear tests have been carried out for monotonic and cyclic loading on

short shear walls. Experimental testing for cyclic reversed loading is needed to further

investigate the effect of the many variables and to perfect the reinforcing details to

assure an optimum strength, stiffness, and ductility of slender shear walls. Important

variables to be considered are the arrangement of reinforcement, details for confinement,

details for splices, effects of strength of concrete, influence of floor level construction

joints, and others. Also, the extent of the yielding region needs to be investigated

experimentally. The studies should encompass rectangular, dumbbell, I-sections, and walls

with openings. Experimental studies would provide moment-rotation and shear-deflection

VI-40
characteristics needed to develop a mathematical model to be used for dynamic response

studies of structures containing shear walls. Also dynamic experimental studies on the

shaking table are desirable to ascertain the reproducibility of the general response characteristics using mathematical models.

In the area of analytical dynamic response studies, work is needed on shear wall and

shear wall-frame type structures to determine the ductility demands on shear walls. Also,

the required stiffness of the structure to respond to various categories of earthquake

ground motions should be investigated to determine the eventual number of shear walls needed

in a building to control its earthquake response.

Conclusion

In conclusion, it should be pointed out that until now safety against collapse has

been the major preoccupation of earthquake engineering. However, we now should enter a

second stage of our development in which, in addition to safety, damage control should be

our major goal. Judging from the behavior of multistory reinforced concrete buildings in

earthquakes, it seems that to achieve damage control the ductile shear wall may be the most

logical solution. Actually, from observations in earthquakes, it seems that we can no

longer afford to build our multistory buildings without shear walls.

*

VI-41
References

1. Fintel, Mark, "Quake Lesson from Managua: Revise Concrete Building Design?', Civil

Engineering - ASCE, V. 43, No. 8, Aug. 1973, pp. 60-63.

2. "Preliminary Report - The Behavior of Reinforced Concrete Structxires in the Caracas,

Venezuela Earthquake of July 29, 1967", Portland Cement Association, Skokie, 1967,

51 pp.

3. Corley, W. Gene, and Hanson, John M. , "Design of Earthquake-Resistant Structural Walls",

Proceedings , Fifth World Conference on Earthquake Engineering (Rome, 1973).

4. Brada, P., "Shear Strength of Low-Rise Walls with Boundary Elements", PhD Thesis, Lehigh

University, Bethlehem, Pa., 1972.

5. Barda, F. , Hanson, John M. , and Corley, W. Gene, "An Investigation of the Design and

Repair of Low-Rise Shear Walls", Proceedings , Fifth World Conference on Earthquake

Engineering (Rome, 1973)

6. Cardenas, A. ex, Hanson John M. , Corley, W. Gene, and Hognestad, Eivind, "Design

Provisions for Shear Walls", ACI Journal Proceedings , V. 70 No. 3, Mar. 1973 pp. 221-230.

7. Cardenas, Alex E., and Magura, Donald D. , "Strength of High-Rise Shear Walls-Rectangular

Cross Sections", Response of Multistory Concrete Structures to Lateral Forces , SP-36,

American Concrete Institute, Detroit, 1973, pp. 119-131.

8. Pauley, T. , "Some Seismic Aspects of Coupled Shear Walls", Proceedings , Fifth World

Conference on Earthquake Engineering (Rome, 1973)

.

9. Slase, E. A. B. , and Fintel, Mark, "Strength, Stiffness and Ductility Properties of

Slender Shear Walls", Proceedings , Fifth World Conference on Earthquake Engineering

(Rome, 1973).

VI-42
Fr»» Frtmt Fr», Will CemblMd Fnm A Wll

T'lg. 1 —Shear wall-frame Interaction

Fig. 2—Two recent Bank Buildings in Managua, representing different structural systems. On the left, Banco

Central and on the right, Banco de America.

Fig. 3 —Banco Central—Plan

VI-43
Fig. A—Banco Central—Inside damage

Fig. 5—Banco de America—PIdn

iiiiiiL(iM.

niiiiiiiKKii

Fig. 6—Insurance Building—Managua Fig. 7—Enaluf Building—Managua

Fig. 8—Social Security Building—Managua

11^"B^

Fig. 9—Typical skeleton of a multistory building in Caracas, Venezuela

VI-44
Fig. 10—Damage fo brittle hollow clay tile infill in Caracas, 1967

Fig. II —The Plaza One Building —Caracas

Fig. 12—Party Headquarters, Skopje,

Yugoslavia

Reference aiis 0

Concrete

Fig. 13 —Olive" View Hospital—San Fernando

WALL ELEVATICM

STRAIN AND STRESS DISTRIBUTION

NONLINEAR BEAM THEORY.

Rg. 14—Nonlinear beam theory assumptions

VI-45
Fig. 15—Load-moment interaction envelope

16—Ductility versus axial load

VI-46
SURVEILLANCE OF CORPS OF ENGINEERS STRUCTURES

IN EARTHQUAKE-PRONE AREAS

by

Keith O. O'Donnell

Office, Chief of Engineers

Washington, D.C.

ABSTRACT

This paper explains the program which the Corps of Engineers has adopted to assess the

effects of earthquake activity concerning the structural behavior of Civil Works hydraulic

structures. The program encompasses reporting earthquake effects and an instrumentation

system on dams and appurtenant structures for monitoring. Post- earthquake inspections will

be conducted to detect significant structural distress and provide information for the

necessary remedial measures for damaged structures. The strong motion instrumentation will

provide a record of ground and structure motion during earthquakes. Data provided from

this program should be an aid in selecting design earthquakes and may provide preliminary

guidance in design procedures for use in predicting, from small earthquakes, the behavior

structures subjected to larger design earthquakes. The accumulated information from the

entire program should help in improving the design criteria for future designs.

Key Words: Earthquakes, Hydraulic Structures; Inspection; Instrumentation; Strong-motion

Accelerographs

.

VI-47
Reporting Earthquake Effects

General

In 1969 the Corps of Engineers established a program for assuring the structural integrity and operational adequacy of major Civil Works structures following the occurrence of

significant earthquakes. This program is primarily concerned with damage surveys following

the occurrences of earthquakes greater than Richter Magnitude 5. The evaluation of structures following an earthquake is of paramount concern to insure structural stability, safety

and operational adequacy for structures whose failure or partial failure would endanger

lives or cause substantial property damage. Structures included in the program are dams,

navigation locks, powerhouses and appurtenant structures which are operated by the Corps

of Engineers. Other facilities such as major levees, flood walls and pxraiping stations

which are designed and constructed by the Corps of Engineers are included although their

maintenance and operation are by other agencies.

Guidelines for Post-Earthquake Inspections

Earthquakes may cause hidden or obvious damage to all types of structures. It is intended that prompt post-earthquake field inspecting and reporting will aid in determining

the appropriate remedial measures for structures that have been damaged. All important Civil

Works structures within a prescribed area that has been subjected to an earthquake of

sufficient magnitude to produce a Modified Mercalli Intensity of 5 (MM5) or greater will be

inspected for damage. This prescribed area is defined as the area enclosed by the Intensity

5 Contour or "isoseismal. " Information to identify the prescribed area usually can be obtained from the Coast and Geodetic Survey or U.S. Geological Survey. However, since the

qualitative or quantitative ground motion data are usually not immediately available, the

prescribed area for post-earthquake inspection will be determined by magnitude scale. For

a Richter Magnitude less than 5.0, no inspection is required unless specfic reports of

possible damage are received from project offices. For a Richter Magnitude 5.0 through 7.0,

inspection of projects within 200 miles of the epicenter is required and for a magnitude

exceeding 7.0 inspection is required within 500 miles of the epicenter.

There are many types of structural damage induced by ground motion from earthquakes

that need to be reported. Any change in appearance or functional capability of a major i Civil Works structures should be evaluated. Examples would include cracked or shifted I bridge pier footings or other concrete structures; turbidity or changed static level of

water wells; cracks in concrete dams or earth embankments; misalignment of hydraulic controlj

structures or gates; loss of freeboard by settlement; development of a localized quick

j condition within an embankment section or new seepage paths through an embankment or founda-j

tion. J

Inspection and Evaluation I

Generally, the structures that would be of most concern following an earthquake are also

the structures which are covered under the periodic inspection program which the Corps of

Engineers adopted in 1965 for major Civil Works structures. Where feasible, the instrumentation and prototype testing program included in the periodic inspection program to monitor

structural performance will serve in the post-earthquake safety evaluation program. AddiVI-48 i
tional special types of instrumentation are incorporated in selected structures in which it

is desirable to measure forces, pressures, loads, stresses, strains, displacements, deflections or other conditions concerning damage and structural safety in case of an earthquake.

A report will be prepared of the post-earthquake inspection of each selected structure

which will include summaries of instrumentation and other observation data, for permanent

record and reference purposes. The U.S. Army Waterways Experiment Station at Vicksburg,

Mississippi is assigned the responsibility for analyzing and interpreting the earthquake

data obtained from the strong motion instrumentation.

Earthquake Aftershock Measurements

A field team has been formed at the Waterways Experiment Station which is capable of

being "mobilized immediately upon notification of the occurrence of an earthquake that

might be expected to produce significant ground motion from aftershocks at any project

operated by the Corps of Engineers. The objective of the field team is to provide a capability for rapid response to earthquake occurrences in recording response of installation,

such as dams, to earthquake aftershocks. Instrumentation will be utilized to supplement

existing instruments which may have been installed as a part of the Corps of Engineers

strong-motion instrumentation program. The measurements obtained will provide approximate

input for analysis of the response of the structures under loading by more severe earthquakes. Additionally, they will provide input for the analysis of other similar structures,

and by comparison of observed and calculated response of the structures, allow evaluation

of existing analytical procedures and aid in developing improved methods for seismic design.

The instruments presently available for this purpose include four Teledyne model RFT-250

accelerographs , which have response ranges from 0.25g to l.Og, and four Kinemetric SMA-1

strong-motion accelerographs with l.Og maximum range, incorporating WWVB time base radio

receivers. These instruments record accelerations on 3-orthogonal axes, and the standard

time base will permit correlation of events that are recorded on the various instruments.

This equipment is maintained on standby so that it is available on a 24 hour basis. The

equipment is packaged so that it may accompany the field team aboard commercial aircraft.

Seismic Instrumentation Program

General

The Corps of Engineers has adopted a policy providing seismic instrumentation for Civil

ji Works structures in regions of significant seismic activity to measure ground motion and

'; response of the structure. Each installation is planned to suit the particular structure,

geologic and seismic conditions. The seismic instrumentation includes strong-motion

''recording accelerographs, peak recording accelerographs, seismoscopes , and hydrodynamic

pressure gages. The types of strong-motion accelerographs used by the Corps of Engineers

j'are shown on Figures 1 and 2. Figure 3 shows the peak recording accelerograph and Figure

'4 the seismoscope. The seismic zone map of the United States for the contiguous states is

;:„shown in Figure 5.

: Seismic Instrumentation for Concrete Dams and Intake Towers

The normal installation for concrete dams 150 feet or over in height in all seismic risk

(Zones 4 or 3 and those in Zones 2 in the western part of the United States consists of a

VI-49
minimum of three strong-motion accelerographs . One accelerograph is located on top of

the higher dam monoliths, one in the upstream gallery near the base of the same monolith

and one in the bedrock foundation at a distance downstream of the toe of about three times

the dam height. For the larger dams (300 feet and over) , or where foundation conditions

warrant, a fourth accelerograph should be located about mid-height of the principal instrumented monolith. For dams 150 feet or more in height in Zones 2, other than in the western

part of the United States, at least one strong-motion accelerograph is located on the bedrock, structure near the dam to record ground motion. Intake gate towers 150 feet or more

in height located in Zones 4 or 3 and in Zone 2 in the western part of the United States are usually instrumented with one accelerograph at the base and one at the top of the tower.

Instruments which are suitable for installation at concrete dams are strong-motion triaxial instruments such as Kinemetrics SMA-1, manufactured by Kinemetrics, Inc., San Gabriel,

California and Teledyne RFT 250 or RFT 350, manufactured by Earth Sciences Division of

Teledyne Corp., Pasadena, California. The accelerographs should have a maximum operating

range up to 0.5g or l.Og. The l.Og instruments would be appropriate at the crest of high

dams and the 0.5g instruments at other locations near or at the ground surface. Seismoscopes

to measure the relative response of different foundation or geologic formations within the

immediate project vicinity are installed to supplement the conventional strong-motion accelerographs located at the foundation or abutments . Hydrodynamic pressure gages are installed

on the upstream face of high dams and intake towers (over 300 feet) to measure the increased

pressure of the reservoir water during earthquake occurrences.

An example of seismic monitoring instrumentation for concrete dams is that of Dworshak

Dam which is shown on the drawing of Figure 6. This dam is located on the North Fork of the

Clearwater River in Idaho and has a maximiim structural height of 717 feet. The instruments

at Dworshak consist of four time recording strong-motion accelerographs, thirteen triaxial

peak accelerographs and six dynamic water pressure cells. These instruments are generally

located within the highest monolith of the dam, in the powerhouse superstructure concrete

walls, around the lake in the vicinity of the dam and at two bridges located approximately

18 and 40 miles upstream from the dam. Three Kinemetric, Type SMA-1 accelerographs are

located in monolith 23 in small instr\ament rooms at elevation 980, 1260 and 1603. The

fourth SMA-1 is located on the right bank 1430 feet downstream of the dam in a steel and

concrete vault. This fourth instrument is to measure ground shock-wave intensity for a

basis of comparison of intensities measured at different locations within the structure.

Six dynamic water pressure cells are located on the upstream face of the dam at elevations

1050, 1160, 1220, 1280, 1350 and 1435 to measure and record water pressure changes against

the structure in the event of an earthquake. Four of the triaxial peak recording accelerographs are located in the powerhouse, three in the thin superstructure walls and one in the

substructure to monitor and compare shock-wave accelerations. Five triaxial peak recording

accelerographs are installed around the lake in the vicinity of the dam to monitor possible

varying intensity of ground shock-waves at different locations. The other four triaxial

accelerographs are located at the two bridges upstream from the dam.

VI- 50 I
Seismic Instrumentation of Earth and Rock-Fill Dams

Earth and rock-fill dams 100 feet high and greater and located in seismic risk zones

2, 3, and 4 are instrumented for strong-motion measurements. All areas west of the Rocky

Mountains are considered at least zone 2 for instrumentation purposes. Dams less than 100

feet high and/or located in zones other than stated above are considered for instrumentation

dependent upon embankment and foundation materials, particularly if such materials are con- sidered susceptible to liquefaction. Locations of instruments for an embankment dam are

shown on Figure 7. All embankment dams in seismic zone 4 and in zones 3 west of the Rocky

Mountains are instrumented in accordance with Category C as depicted in Figure 7 except in

special cases where more complete measurements are desired and in these cases Category D

is followed. Most other structures are instrumented in accordance with Category B. The strong-motion accelerographs , such as the Teledyne RFT-250 or 350 and Kinemetric

SMA-1 are also recommended for embankment dams. Because of difficulties with transmission

of signals from transducers to remote recorders, central recording systems are not generally

recommended. Seismoscopes are used on different geologic deposits in the immediate vicinity

of the dams to provide an estimate of the influence of geologic conditions on ground motion

intensity and to supplement ground motion information obtained from the accelerographs.

However, seismoscopes are not installed on embankment dams because of the difficutly in

interpreting the records from this type of installation.

Seismic Hazards of Reservoirs and Landslides

Experiences elsewhere in the world have indicated that the filling of large reservoirs

might induce earthquake activity. An example of this is the Koyna Reservoir in India. The

Corps of Engineers has designed and built a number of high dams with large reservoirs and

no earthquake activity has been reported as a result of reservoir filling. The Corps' high

dams are as follows: 34 dams over 200 feet high, 15 dams over 300 feet high, 6 dams over

400 feet high and 2 dams over 600 feet high. Of these dams 24 have a reservoir volume

larger than one million acre feet. Special microseismic instrumentation to measure

reservoir effects has been installed at several projects.

The importance of possible landslides into the reservoir cannot be overlooked. While

not caused from earthquake activity, the Corps of Engineers had an experience at Libby Dam

when during construction a small slide occurred in bedrock at the left abutment which

damaged the concrete plant. As a result a comprehensive study was made to determine

possible effects should a geologically similar landslide into the reservoir develop. A

model was built at the Waterways Experiment Station to measure the effects of such a landslide. To increase stability and minimize concern at Libby Dam, a massive rockfill buttress was constructed along the upstream abutment slope. Extensive instrumentation was

developed to detect and measure changes in subsurface conditions indicating potential

instabilities. The instrxaments included the following: extensometers , porepressure

measuring devices, inclinometers, and micro acoustic noise transducers.

VI-51
Strong-Motion Instrumentation Installations

As of July 1974, 38 dam projects had been instrumented with 117 strong-motion accelerographs, 52 seismoscopes and 18 peak recording accelerographs . Additional instrumentation

is currently planned for 56 projects which includes 171 strong-motion accelerographs, 50

seismoscopes and 17 peak recording accelerographs. The status of the instrumentation program is shown in Table 1.

Seismic Engineering Branch of the U.S. Geological Survey (USGS) is responsible for I maintenance of the strong-motion instruments installed on Corps of Engineers structures. I

The record of ground and structure motion during earthquakes should provide valuable information to form a basis for selecting design earthquakes and for establishing the validity of

dynamic design procedures. Also, the record of small earthquakes will be helpful in making

preliminary estimates of the behavior of the structures during larger design earthquakes.

The instrumentation program is of great importance to the Corps of Engineers because

of the many large reservoir projects for which the Corps is responsible. It is hoped that

should significant earthquakes occur, the results of seismic instrumentation will give a better understanding of dynamic response of structures and perhaps lead to improved designs

for seismic behavior.

VI-52
References

1. Engineer Regulation ER 1110-2-1802, 10 September 1969, "Reporting Earthquake Effects",

Corps of Engineers, U.S. Army.

2. Engineer Technical Letter ETL 110-2-130, 30 July 1971, "Seismic Instrvimentation of

Concrete Dams and Intake Towers", Corps of Engineers, U.S. Army.

3. Engineer Regulation ER 1110-2-103, 9 August 1974, "Strong Motion Instruments for Recording Earthquake Motions on Dams," Corps of Engineers, U.S. Army.

4. Engineer Technical Letter ETL 110-2-195, 9 September 1974, "Instrumentation for

Measurement of Earthquake Motion and Instrument Shelters for Earth and Rock-Fill Dams,"

Corps of Engineer, U.S. Army.

5. Engineer Circular EC 110-2-151, 26 September 1974, "Status Report of CE Strong Motion

Instrumentation Program for Measurement of Earthquake Motions," Corps of Engineers,

U.S. Army

.

I
Figure 1 RFT-350 strong motion accelerograph

.

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VI-54
VI-55
VI-56 i
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VI-57
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VI-63
SCHOOL AND HOSPITAL CONSTRUCTION IN CALIFORNIA

by

John . F . Meehan

Research Director, Supervising Structural Engineer

Structural Safety Section

Office of Architecture and Construction

Department of General Services

State of California

ABSTRACT

The California State Legislature has adopted statutes concerning the regulation of the

design and construction of public school buildings and hospitals. These statutes were brought about because of the rather poor performance of these types of buildings in California earthquakes. Events leading up to these statutes and the methods of their enforcement will be discussed.

Key Words: Building Codes; Building Regulations; Design Criteria; Earthquakes; Hospitals;

Schools.

VI-64
Public Schools

The 1933 Long Beach, California earthquake produced great damage to public school

buildings throughout the devastated area. It occurred at 5:54 p.m. on Friday, March 10,

1933 and, as usual during this time on a Friday evening, there were essentially no students

or teachers in the buildings. Had the earthquake occurred during school hours the number

of deaths to the school children would have been horrifying. The California State Legislature requires school attendance of the youngsters and recognizing that earthquakes may

occur at a time of full occupancy of the children, took measures to prepare and adopt the

Field Act within a month after the earthquake. The rapidity of the passage of this legislation clearly indicates the gravity and concern of the legislature and the public. Each

earthquake that has occurred in California since 1933 has proven the wisdom of this Act.

Reports of earthquake damage in California usually point out the good performance of postField Act school building construction. (1, 2, 3, 4, 5, 6, 7, & 8)

Field Act and Enforcement

The Field Act will be briefly explained and the method of its enforcement by the

Structural Safety Section (SSS) of the Office of Architecture and Construction (OAC) in the

Department of General Services (DCS) will be described. The Field Act is given in Section

3 of Title 21, California Administrative Code (CAC)

.

The Act applies to all new public school buildings and to all additions or alterations

of schools costing more than $10,000. The plans and specifications must be preapred by

registered structural engineers and/or architects in the private sector. The complete con- struction documents including the structural design calculations, geologic reports,

application and fee are siabmitted to SSS. The prints of the plans and specifications are

reviewed (checked) and comments are marked on them by registered structural engineers in

the SSS for conformance with seismic design regulations. To accomplish this, the structural

plans and the structural or anchorage aspects in the architectural, mechanical and electrical plans are rather closely checked. The plans and specifications are returned to the

structural engineer or architect for correction. The tracings and marked documents are

then returned to SSS and compared (back checked) by the SSS representative working together

with the responsible structural engineer and/or architect. Upon mutual agreement of the

revisions, the tracings are stamped approved. A corrected set of approved stamped documents are filed with SSS. Competitive bids are taken and the contract may then be let by

the school district to the lowest responsible bidder. All change orders and addenda must

also be signed by the architect or structural engineer and approved by SSS. Continuous

on-site inspection of all phases of the construction is required by the Field Act and is

provided by a construction inspector approved by the structural engineer and/or architect

in responsible charge of construction, the school board and the SSS. Materials of construction to be tested and special inspections to be performed are established at the time

of back check. Reports are filed by private sector testing laboratories to assure

material conformance with the requirements of the approved documents . The structural

engineer and/or architect in charge of construction are required to supervise or observe

the work of construction. Certificates verifying that the construction fully complies

VI-65
with all of the requirements of the approved documents are filed by the structural engineer,

architect, construction inspectors, testing laboratories and contractor. Making a false

statement is a felony. Field personnel from SSS also visit the projects periodically to

review the construction for possible design and construction errors . A fee of one-half of

one percent of the construction cost is paid by the school district to the SSS to fund this

operation.

It has been frequently expressed by structural engineers in the private sector that the reasons for the good performance of public school buildings in earthquakes are because the

designs are prepared by capable personnel; the independent review of the plan details by

structural engineers equally capable as those who originally designed the structure to assure code compliance and design concept; the inspection provided by the responsible designers; the continuous detail inspection provided by the approved construction inspectors;

and the review of the construction by the SSS structural engineers for conformance of design

principles. In general it is required to pay close attention to proper detailing of the

structure and provide complete and logical means of resisting lateral loads giving full con- sideration to deflections and displacements.

The Field Act also established that regulations may be adopted pursuant to the legislation. These regulations, adopted by SSS, are those of the 1973 Uniform Building Code

together with certain nominal additions and deletions found necessary to accomplish the requirements of the Field Act. These amendments are printed in Title 21, CAC. The lateral

force provisions are basically those recommended by the Seismology Committee of the

Structural Engineers Association of California. A copy of Title 21, CAC, can be obtained

from State of California, Documents Section, P.O. Box 20191, Sacramento, Calif. 95820.

Title 21, California Administrative Code (CAC) and Uniform Building Code (UBC)

Exceptions

The regulations adopted to administer the Field Act are given in Groups 1 and 2 of

Title 21, CAC. The detail technical regulations are given in Group 3, Article 23 through

Article 47. All of the items given in Title 21, CAC in article 23 through 47 are exceptions

or additions to the 1973 UBC. Just a few of these will be mentioned here.

Loading

The minimum acceptable base shear KC factor for any one and two story building is 0.10

as mentioned in Section T21-2316(d) (1) and all frames must meet the ductile requirements.

This base shear factor seldom is applied, as most designs proposed are based upon a base

shear of 0.133, and further, the framing is usually of shear wall design concept.

Any single mass structure on a single column or of a type where a single column must

provide the total lateral resistance, the lateral force base shear KC factor must be 0.30,

and the column must meet the ductile requirements.

Also if building frames are designed for a base shear KC factor of 0.30 or more, the

frame need not be ductile but the colxmns must be ductile. Most frames are designed under

this basis.

In the near future Title 21, CAC will probably be revised to require a KC base shear

factor of 1.5 5imes the code force when the building is located near established active fad.t£

VI-66
Drift is controlled in Section T21-2307(c). This section requires the maximum drift

of 1/16" per foot of height. The drift is also limited to 1/32" per foot of height of

opening where glass is installed in the walls which may deflect in the plan of the wall and

the glass is not separated from the frame such that any in-plan deflection will transfer

load into the glass. In tests, reported by this author at the Second World Conference on Earthquake Engineering in Tokyo, it was shown that glass fails at about 1/16" per foot of

height when bedded in stiff mastic and the glass frame cannot rotate. Wall defelction limitations applied perpendicular to the wall is also given in Section T21-2307(c).

There are diaphragm span/depth ratios given in Table T21-23L. These ratios are primarily to control deflection and stress in the horizontal diaphragms which, in effect, also

control the flexibility of the horizontal diaphragms . Controls are also provided for vertical diaphragms in this table.

Masonry

All masonry construction requires inspection by a specially approved masonry inspector.

His duties are described in Section 2401 of Title 21, CAC. This section also requires cores

of the completed walls to be taken and tested in shear and compression. Cores not only provide an indication of the strength of the materials but also provide an excellent means of

determining the quality of workmanship within the wall.

The allowable stresses in masonry are given in Article 24, but under most designs the

stresses in masonry seldom control the design. All masonry must be reinforced. Maximum

spacing of reinforcing is limited to 24 inches each way in walls and extra ties are required at the ends of all columns and around bolts. Such reinforcing greatly increases

the toughness of the wall. High and low lift grouting is allowed but it is imperative that

it be reconsolidated at the proper time and again after the next lift of grout is placed,

also it is strongly recommended that an expansive type admixture is used in the grout.

Both of these conditions, the reconsolidation and the expansive admisture, greatly reduces

the amount of internal shrinking cracks in the grout. Unreinforced masonry or cavity wall

construction is not allowed.

Methods of rehabilitating existing masonry construction is given in Section T2102422.

Wood

Article 25 of Title 21, CAC coverts the requirements for wood. The exceptions are

rather nominal. The maximum adjustment for stress for duration of loading is limited to 15%

increase. The allowable nail stresses are higher than those given in the current UBC because of the success of designs provided iinder Title 21, CAC. As an example, the allowable

shear load on an 8d common nail in Title 21, CAC is 100 pounds whereas the allowable is 78

pound in the 1973 UBC. These nail stresses have been used successfully since the Field

Act became effective in 1933.

Glue lam construction. Section T21-2511, basically conforms to current industry

standards except the maximum moisture content at time of gluing relates to the expected

moisture content in the area when the building is situated. That is, the maximum allowable

moisture content along the sea coast is 16% whereas in the desert areas it is limited to

10%. Continuous inspection by a specially approved inspector is required for glue lab

VI-67
beam construction.

Unblocked plywood diaphragm allowable loads are a bit lower than that permitted by UBC

as given in Section T21-2514.

Rather complete details are usually given on the plans showing wood framing, openings

in floors and walls, notching, etc. Also careful attention has always been paid to property

anchor wood roofs and floors to masonry or concrete walls.

Concrete

Much of the material in Title 21, CAC in the concrete section refers to proportions,

mixing and placement. There are several additional design and construction requirements

such as those given in Section T21-2406 which call attention to the requirements for clean

construction joint; additional column ties are required by Section T21-2607 on the ends of

all columns; the column ties shall terminate in a 135° bend with 6" extension; Section T21-

2614 requires height/thickness ratio of bearing walls to be not more than 30 and not more

than 36 for non-bearing walls, precast walls require special trim wall reinforcing, anchorage and other requirements; Section T21-2618 requires grouting of all post-tension tendons

anchored with friction type anchors; and Section T21-2621 contains controls on pneumatically

placed concrete.

Steel and Iron

Article 27 covers steel and iron. There is little difference between the basic UBC

requirements and Title 21, CAC. There are, however, controls on steel deck used as diaphragms and load test requirements of steel joists if analyses cannot be readily performed.

Welding inspection is required by capable inspectors who must use all means necessary to assure himself of the proper quality of the weld.

Foundations

Under foundations , the additional special requirements are for inspection of engineered

fill, piles and other miscellaneous items.

Veneer

All veneer over 3/8" in thickness must be anchored by mechanical means. Details of

anchorage and reinforcing of the masonry is given in Article 30. Walls to which veneer is

applied must have stiffness of L/600. The connection is usually by means of sheet metal

dove-tail anchors. The anchor slot is cast into the concrete walls or nailed with lOd

common nails at 12" oc. A nail placed vertically between the end of the anchor and the

back of the slot to take up any slack in the connection. A number 9 wire is placed horizontally in the mortar joint in each horizontal joint containing an anchor. The maximum

spacing of anchors is about 16" oc horizontally and 12" vertically.

Although currently not required by code, it is this authors preference to omit any

veneer, or glass over exit doors.

Roofing

Article 32 provides for anchoring the roofing to the structure including tile. The

anchorage of each tile by means of non-corrodible wires or nails is required to prevent the

tile from sliding off the roof and falling to the ground during earthquake motion.

VI-68
Ceilings

Plaster ceilings must be wire-tied to the ceiling framing to prevent the ceiling from

falling. It was learned that ceiling fastened to wood framing only by means of nails in

withdrawal will allow the ceiling to fall in one single large piece somewhat like a blanket.

Complete details for all types of ceilings and light fixture anchorages must be shown on the

drawing.

Anchorage

Items such as boilers, water tanks, fixtures, cabinets, shelves, window sash, ceilings,

heaters, etc. must be properly anchored to the structure and are shown on the approved detail drawings

.

Field Act Advisory Board

The SSS has an Advisory Board composed of four structural engineers, four architects,

a mechanical engineer and an electrical engineer from the private sector who serve without

pay. Representatives of various state agencies are ex-officio members. This board provides advice on technical and administrative matters relating to the regulations and en- forcement of the Field Act.

Other Legislation

When the Field Act was first adopted in 1933 it related only to new construction and

it remains so today. Existing buildings were not mentioned. Legislation was later adopted

in the Garrison Act which provided that, if a building was examined and found unsafe and the

building was allowed to oontLnue in operation without correction and without attempting to

obtain funds tor the correction if no funds were available, the school board members were individually liable. Many old school buildings were, therefore, never examined to determine

whether they were unsafe. In 1967 legislation was adopted to require a structural examination of all school buildings constructed prior to 1933. In 1968 the statutes were changed

and provided that if found unsafe and not corrected, such buildings could not be used for

school purposes after June 30, 1975. In 1975 legislation was added to provide that if definite steps were taken by June 30, 1970 to correct the deficiency, the building could be

used until June 30, 1977.

When the Field Act was first adopted there was no control established concerning the

location of the school site. As a result, many school buildings were located very close to

active faults. In 1967 legislation was enacted to require a geologic hazards report and to

prohibit siting new schools on active faults or on other geologic hazards. It was later

amended to also require a geologic hazards report for building additions on existing sites.

Currently a geologic hazards report is required when deemed necessary by SSS.

Hospitals

Hospital buildings have frequently been damaged from the larger earthquakes in California and therefore were evacuated just at a time when they were most critically needed to

serve the victoms from other damaged man-made structures.

Following the 1971 San Fernando earthquake, where 50 deaths or 85% of the total deaths

of patients or employees resulted from hospital building collapse or inoperative equipment,

where four hospitals were evacuated, and where 17 hospitals were damaged, legislation was

VI-69
passed by the state legislature in 1972 requiring construction approval procedures similar

to those in the Field Act to apply to hospitals. There are some differences.

The Department of Health (DH) administers the act, but is required to contract with the

DGS to perform a similar service on hospitals as they do on public school buildings. It

requires that all plans and specifications shall be prepared under the responsible charge of

a California registered architect or structural enigneer or both; it requires that a structural engineer shall sign plans and specifications related thereto; it requires a geologic

hazards report on all but the small buildings; it requires the geologic data to be reviewed

by an engineering geologist and the structural design to be reviewed by a structural engineer;

it requires a fee based upon the estimated cost and a further fee based upon the final cost;

it requires continuous on-site inspection by competent personnel and administration of the

work of construction to be under the responsible charge of such structural engineer or architect; it requires signed certificates indicating conformance with the approved documents from all those involved in the construction; it authorizes the DH to adopt regulations with the advice of the DGS to carry out the desires of the bill; and it establishes

a Building Safety Board to advise and act as an appeals board with regard to seismic safety;

and it established that any person who violates the provisions are guilty of a misdemeanor.

This bill is Chapter 1130 of the 1972 Statutes and is given in Division 12.5 of the Health

and Safety Code.

The SSS enforcement of this act is quite similar to the procedures followed under the

Field Act and will not be repeated here.

Building Regulations

The legislation mandates that ".. .hospitals .. .be completely functional to perform all

necessary services ... to resist, insofar as practicable, the forces generated by earthquakes,

gravity and winds..." places a new direction in building codes. Recognizing this and the

need for immediate input from the private sector, DH contracted with a structural engineer,

a mechanical/electrical engineer and an architect as consultants to SSS to prepare the regulations. They were adopted as emergency measures, then public hearings were held when

input was provided by the Seismology Committee of the Structural Engineers Association of

California and other interested groups.

The building regulations for California hospitals are given in Title 17, CAC and contains the special provisions for the seismic load levels and performance. The remainder of

the regulations are essentially those required for public school buildings. As mentioned

previously, these latter regulations consist of the 1973 Uniform Building Code as the basic

code, together with nominal additions and deletions found necessary to accomplish the desired results. A copy of Article 23, General Design Requirements can be obtained from the

State of California. The remaining regulations are essentially those given in Title 21,

CAC.

A few of the highlights for earthquake forces of Article 23, General Design Requirements

in Title 17, CAC will be presented. There are two basic methods of design. One may be by

dynamic analysis. Method A, and the other by static analysis. Method B. Section T17-2314 of

Title 17, CAC requires that buildings over 150 feet in height or those with highly irregular

VI-70
shapes, or other unusual structural features must have a Method A or a dynamic analysis.

This requires that the structural and deflection capacities shall be sufficient to resist

the effects of earthquakes upon the structure as deteinnined by dynamic analyses. These

analyses shall be based upon the ground motion prescribed for the site in a geotechnic report. The report shall consider the seismic event that may be postualted with a reasonable

confidence level within a 100 year period.

The analyses may be based on appropriate time-histories or response spectra with

percentages of critical damping consistent with the strain levels in the structural materials.

All natural modes of vibration with periods greater than 0.05 seconds shall be considered in

the analyses. If the stresses calculated by elastic analyses exceed the nominal yields

stress of the structural materials, the structural elements to be used shall be justified

by approved methods which reconcile the equivalent inelastic deformations with the ductility

of the structural elem.ents.

The base shear resulting from the dynamic analyses shall be not less than 80% of the

base shear calculated from the provisions of Method B. If the base shear as determined by

dynamic analyses must be increased to meet this requirement, the design spectra shall be

normalized to a proportionately higher value.

Under Method B, the static analysis, the lateral load level has been raised above that

provided for other occupancies given in the Uniform Building Code. The base shear equation

is

V = KCW

where K shall be 3.00 for all buildings, tha value of C is the same as that given in the

1973 Uniform Building Code, except that the combined value of KC shall be 0.25 for all and

two story buildings and need not exceed 0.25 for any building. The method for calculating

the period, T, and the load distribution is the same as that of the 197 3 Uniform Building

Code. Since all of California is located in Zone 3 of the Uniform Building Code, zone

factors are not mentioned in these regulations and tha importance factor currently being

considered for inclusion in the Uniform Building Code is included in the KC value. All

lateral force resisting frames must be ductile; however, in one and two story buildings

when the KC value used in the design is 0.50 or more, the frames need not be ductile but the

columns must be ductile. The C table is more extensive than that given in the Uniform

P

Building Code and the load level is higher in all cases. Anchorage requirements for such

items, equipment, elevators, piping, cabinets, lights, ceilings, partitions, etc. are

presented.

Building Safety Board

The states establish the formation of a Building Safety Board to advise the DH with

regard to seismic strautural safety and to act as a board of appeals in the enforcement of

the Act. This board is composed of members qualified by close connection with hospital design and construction and are highly knowledgeable in their respective fields with particular reference to seismic safety. The board consists of eleven members of which there are

two structural engineers, two architects, one engineering geologist, one soils engineer and

one hospital administrator appointed by the Director of Public Health from nominations of

VI-71
technical and professional associations. There are also sdx ex-officio meinbers from state

offices

.

This board has provided a great deal of input to the Department of Health concerning

the operation of this Act. It reviews the regulations to be adopted, has studied and recom- mended methods of approving work in existing buildings and advises the Department as requested.

The board has established standing sub-committees on structural, architectural, mechanical and eletrical, geotechnic and hospital operations. These committees act as liaison

between the board and the respective technical associations and provide advice on assigned

subjects. There are also ad hoc committees on appeals procedure, plan checking fees, board

rules and procedures, and legislative intent. A board member is the chairman of each s\±icommittee and ad hoc committee.

VI-72
References

1. Steinbrugge, K. V. and Moran, D. F. , An Engineering Study of the Southern California

Earthquake of July 21, 1952 and of Its Aftershocks. Bulletin of the Seismological

Society of America, April, 1954, p. 344 and p. 255.

2. Steinbrugge, K. V., Cloud, W. K. , and Scott, N. H. , The Santa Rosa California Earthquake

of October 1, 1969, Coast and Geodetic Survey, 1970, p. 38 and p. 10.

3. Steinbrugge, K.V. , Schader, E. E. , Bigglestone, H. C. , and Weers, C. A., San Francisco

Earthquake of February 9, 1971, Pacific Fire Rating Bureau, p. 2.

4. Meehan, J. F., Degenkolb, H. J., Moran, D. F., Steinbrugge, K. V., Cluff, L. S., Carver,

G. A., Matthiesen, R. B., and Knudson, C. F., Managua Nicaragua Earthquake of December

23, 1972, Reconnaissance Report, Earthquake Engineering Research Institute, 1973, p. 167

5. National Academy of Sciences, National Academy of Engineering, The San Fernando Earthquake of February 9, 1971, p. 13.

6. Jennings, P. C, Engineering Features of the San Fernando Earthquake, California

Institute of Technology EERC 71-02, p. 276.

7. National Bureau of Standards, Engineering Aspects of the 1971 San Fernando Earthquake

Building Science Series 40, 1971, p. 311.

8. Steinbrugge, K. V., Schader, E. E. , Earthquake Damage and Related Statistics, San

Feranado Eqrthquake February 0, 1971, U. S. Department of Commerce, 1973, Vol. I,

Part B, p. 713.

VI-73
IMPROVED EARTHQUAKE RESISTIVE DESIGN AND

CONSTRUCTION OF SINGLE-FAMILY RESIDENTIAL DWELLINGS

by

G. Robert Fuller, P.E.

Office of Underwriting Standards

Housing Production and Mortgage Credit

Federal Housing Administration

U.S. Department of Housing and Urban Development

ABSTRACT

This paper presents results of the research program being carried out by the Applied

Technology Council (ATC) of the Structural Engineers Association of California (SEAOC) under

the sponsorship of the U.S. Department of Housing and Urban Development. The objective of

this project was to review and evaluate available manuals, literature and standards con- cerning design and construction of residential dwellings and response of such structures

to earthquake motions; and to then develop a manual of recommended practice for earthquake

resistive design and construction. This manual would be primarily directed toward builders,

building officials, field inspectors, and house designers and would contain recommended

construction details, architectural layouts, design recommendations and types of construction recommended or to be avoided. The manual is now in draft form and copies are available

for review. A synopsis of research results developed to date will be presented in this

paper. Code comparisons, problem areas, tentative recommendations will be discussed.

The research contract with ATC is not due for completion until June 30, 1975, therefore

this paper is an interim report. The final report with the completed manual will be distributed to members of the U.S. - Japan Panel on Wind and Seismic Effects, when available.

Key Words: Buildings; Building Codes; Construction; Dwellings; Earthquakes; Seismic Design

Criteria.

VI-74
Introduction

The U.S. Department of Housing and Urban Development (HUD) awarded a one year contract

to the Applied Technology Council (ATC) of the Structural Engineers Association of California, on June 30, 1974. The primary objective of this research is to develop a "Manual of

Recommended Construction Practice for Earthquake Resistive Residential Dwellings" to be

used by builders, building officials, field inspectors and house designers.

The manual is to explain the structural behavior of single-family and townhouse

residential construction subjected to forces produced by earthquake shocks. HUD Minimum

Property Standards and other building code requirements will be illustrated. Sound practical construction methods and details for the reduction of structural damage due to earthquakes, will be covered. The manual will also illustrate recommended construction details,

architectural layouts, types of construction recommended, or to be avoided, and methods of

installing mechanical equipment to resist seismic forces.

Scope of Manual

Tasks

Several tasks make up the total project:

Task 1: Review of Literature ; Review literature and reports on damage to residential

structures as a result of recent earthquakes. An analysis and bibliography of

literature reviewed outlining the engineering causes for identified damage to buildings

and outlining possible solutions and corrective actions was required.

Task 2; Review HUD Minimum Property Standards (MPS) and Building Code Requirements:

Review the HUD-MPS, Manual of Acceptable Practice and appropriate building codes to

determine their adequacy for preventing earthquake damage to single-family residences.

Recommend improvements in terms of structural performance based on reviewed reports

of past earthquakes.

Task 3: Develop Construction Details: Develop typical engineering drawings,

illustrations and details which are required to resist earthquake forces with appropriate descriptions and explanations to be easily understood by non-technical

representatives of the housing industry. Details are to be categorized, as appropriate, by Seismic Zones 1, 2, and 3 as delineated in the Uniform Building Code, "Seismic

Risk Map of the United States." Illustrations shall include the types of construction

most commonly used for residential dwellings in earthquake prone areas of the country,

i.e.: wood frame with siding, wood frame and brick veneer, wood frame and stucco,

brick or concrete block masonry, steel or aluminum frame with siding or masonry veneer

and other prevalent conventional combinations of framing, materials, and building

components

.

Construction details will include those for basement and slab foundations, dwelling

structures, utility and mechanical equipment installation, chimneys and fire places,

attached garages, and other architectural and structural components which may affect

the strength, rigidity or stability of dwellings.

Task 4: Prepare Manual and Supplementary Engineering Analysis Report ;

VI-75
Task 4: Prepare Manual and Supplementary Engineering Analysis Report: To include the

following general subjects:

1. Introduction

2. Commentary on principles of seismic force design.

3. Illustrative typical construction details. Description and discussions of details,

as developed in Task 3.

4. Recommendations:

a. Concerning plan checking

b. Concerning construction inspection at site

5. Summary, general conclusions and recommendations

6. Supplementary engineering analysis report including discussion of theory and design

calculations to justify typical details developed to resist seismic forces.

Task 5: Presentation Material : Presentations to include 35mm slides, text material

and voice tape which will be used to explain contents of the manual to user groups -

builders, building officials and designers.

Task 6: Presentations to User Groups : Four presentations to home builder organizations

with invitations to designers and building officials to acquaint them with contents of

manual. Presentations will be conducted in Los Angeles, San Francisco, Portland and

Seattle at different times.

Major progess has been achieved on all of the tasks preliminary to the final presentations. Following is a general synopsis of accomplishments to date including philosophical

discussions by the subcontractor to ATC: Ralph W. Goers and Associates of Los Angeles,

California.

General Philosophy

Economic considerations must be reviewed not only in terms of life/safety but also

damage control. Single-family dwellings always perform well in terms of life/safety.

However, since HUD is an insuring agency, damage control should also be a factor. There is

no doubt that if more shear resistance were provided in residences, damage figures would be

considerably lessened. The approach to be used in the Manual should be that houses, like

other structures, should conform to design requirements of the Uniform Building Code (IBCi

Arbitrary requirements and exceptions in the UBC for single-family framed structures should

not be permitted unless justified by a thorough structural analysis covered under this research project. It is the intent to minimize the degree of damage rather to prevent all

damage

.

This approach will either limit design or more probably add some additional cost to

home construction. There is little doubt that the additional cost is the least expensive

earthquake insurance available. However, the final decision will be up to HUD after re- viewing all of the applicable data concerning earthquake probabilty, intensity and frequency and if justified by reference literature, field observations or calculations.

Finally, governmental agencies have normally led the way in developing both social

and structural changes in our system. It seems only logical that requirements should exceed

building code minimums where such can be justified by reference literature, field observations or calculations.

VI-76
Seismic co efficients for single- family construction are not specified in some jurisdictions and also vary from code to code. It is felt that the UBC value of 0.133g for Zone

3 is sufficient. This would coincide with field observations of damage and with current

HUD-MPS requirements. This conclusion is felt to be justified by the following:

1. One-story homes performed reasonably well in San Fernando earthquake and they

probably would have performed even better if built according to the proposed

manual

.

2. Factors of safety for plywood diaphragms are consistently at or above 3.0. Therefore, houses properly designed and constructed using plywood shear panels to re- sist 0.133g would actually be capable of resisting 0.40g ultimate load.

Virtually all homes built recently have wood framed second floors and roofs. Current

engineering analysis generally assumes the wood diaphragm to be flexible, therefore lateral

loads are distributed to shear walls on a tributary area basis. Field observations indicate

that this assumption is not necessarily correct. Preliminary calculations show that the

second floor diaphragm is approximately ten times as rigid as the longest transverse shear

wall. Relative rigidities of shear resisting elements used in single-family framed con- struction are difficult to determine. General conclusions, however, can be reached on how shear resisting elements act in relation to small wood disphragms

.

Analysis reveals that shears are generated in the first story walls of a small twostory house in excess of allowable values for any of the shear-resisting materials required

by Section 2518(f) 5 of the UBC. Proposed requirements will either result in plywood shear

panels or much longer panels of material of lesser shear resistance for two-story and splitlevel homes.

A split-level house plan was analyzed by use of tributary area method. This typical

type of house received severe damage in one case and collapsed in two other cases during the

San Fernando earthquake. Analysis revealed that the wall between the garage and family room would receive around three-quarters of the total lateral load. If the diaphragm were

analyzed as a cantilever beam in houses of this type, the interior wall would receive close

to 100% of the lateral load. It was found that one of the greatest deficiencies of wood

frame construction is its lack of resistance to torsional racking. In the one case examined

the floor diaphragm rotated in plan, lifting the rear wall from its foundation and leaving

only the center wall to resist further earthquake motions. It is obvious from these previous examinations that analysis techniques must be evaluated and perhaps be revised.

Collection of Data

Several interviews have been completed or arranged with architects, engineers, HUD officials, building department officials and building contractors in several areas of the U.S.

A questionnaire has also been developed for mailing to all HUD Field Offices to determine

types of single-family construction most prevalent in the U.S. (see Attachment No. 1).

Construction methods will also be determined.

Preliminary information indicates that in earthquake prone areas of the U.S., over 90%

of all houses presently being constructed are wood frame with either stucco, wood siding or masonry veneer as an exterior finish and either plaster or gypsum drywall on the interior.

VI-77
Data is being collected on shear wall diaphragm properties for the multitude used:

plywood, stucco, lath and plaster, gypsum wall board (drywall) , fiberboard and others.

Organization of Manual

The following is a general outline of the proposed manual:

Introduction

Acknowledgements

Chapter I - Purpose and Use of Manual

1. Audience

2. Economics vs. Safety

3. Definitions

4. Geographical Scope (Seismic Zone Map)

5. Observations of Earthquake Damage

6. Use of Manual

Chapter II - Principles of Seismic Design

1. Inertial Movement

2. Walls as Vertical Beams

3. Floor and Roof Diaphragms

4. Shear Wall Diaphragms

5. Shear Resistance of Materials (Graphs & Tables)

a. Roofs

b. Floors

c. Walls

6. Chords and Struts

a. Splices and Connections

7. Load to Shear Wall Calculations

8. Effects of Deformation

a. Glazing

b. Other brittle materials

9. Fireplace and Chimney Design and Anchorage

10. Mechanical Equipment

11. Other Non-Structural Items

12. Basement Construction

13. Layout Configurations

14. Glossary of Terms

Chapter III - Construction Details

1. Shear Wall Design and Details

a. Ho Id- Down Anchors

b. Sill-Bolt Size and Spacing

2. Shear Transfer Details

3. Struts and Chords

4. Connection of Studs to Sill Plates

VI-78
5. Exterior Corner Framing and Anchors

6. Fireplaces

7 . Mechanical Equipment

8. Other Non-Structural Items

9. Masonry Construction

10. Steel Frame Houses

Wind design will be covered to the extent that comparisons can be made between wind

and earthquake forces for the entire structxare not for individual elements. Seismic Zones

1 and 2 in the U.S. will be covered together since wind forces generally govern. Seismic

Zone 3 will be covered in a separate section. Even though wind could govern in some areas,

the dynamic effort of earthquakes precludes a comparison with wind forces

.

Review of Building Codes and HUD Minimxjm Property Standards

Unifom Building Code (1973 Edition)

1. Sec. 2308 Wind Pressure : Buildings shall be designed for wind pressures indicated

in tables and maps. Uplift wind pressures for enclosed buildings are 3/4 of the

uniform wind pressures specified and for unenclosed buildings are 1 1/4 times those

specified. Overturning moments calculated shall not exceed 2/3 of dead load re- sisting moment.

Based on experience in Phoenix and Los Angeles, failures due to wind are generally caused by uplift on the roof or to walls by inadequate anchorage of walls

to roof or other supporting diaphragms. The conclusion drawn from observing wind

damaged homes is that either the factor for uplift on iinenclosed structures or the

wind zone map is incorrect. Horizontal wind pressures as applied to the whole

structure appear to be too high.

Despite the fact that uplift forces seem to be critical, overturning requirements for shear walls are questionable since shear forces appear to result in

answers which are too high.

2. Sec. 2313 Anchorage : Masonry and concrete walls shall be anchored to all floors

and roofs to resist a minimum force of 200 plf of wall or the calculated horizontal force, whichever is greater; continuous crossties shall be provided in roof

and floor systems between diaphragm chords.

The force requirement of 200 plf appears to be too low. In many areas of the

country this anchorage requirement and the diaphragm crossties are virtually

ignored. This has proved disastrous not only for earthquake loads but also for

wind forces as well. The manual will provide for minimum sill bolts and horizontal connections to roof framing, roof sheathing of blocking between roof framing

members

.

3. Sec. 2314 Earthquake Regulations : This section defines various systems utilized

to resist earthquakes, formulas to determine equivalent static forces for design,

distribution of horizontal loads, overturning, setback requirements, etc.

As applied to residential construction, this section specified that virtually

all residences are "box systems" without vertical load carrying space frames.

VI-79
Lateral forces are resisted by shear walls or braced frames and the force to be

resisted in 0.133 times gravity loads.

4. Sec. 2514 Wood Diaphragms: Connections and anchorages capable of resisting design

forces shall be provided between diaphragms and resisting elements. In buildings

of wood frame construction where rotation is provided for, depth of diaphragm nor- mal to open side shall not exceed 25 feet nor 2/3 diaphragm width, whichever is

smaller. Straight sheathing shall not be permitted to resist shears in diaphragms

acting in rotation.

One-story, wood-framed structures with depth normal to open side not greater

than 25 feet may have depth equal to width. Where calculations show that

diaphragm deflections can be tolerated, depth normal to open end may be increased

to a depth to width ratio not greater than 1 1/2:1 for diagonal sheathing or 2:1

for special diagonally sheathed or plywood diaphragms.

In masonry or concrete buildings , lumber and plywood diaphragms shall not be

considered to transmit lateral forces by rotation.

The provisions for rotation as presently written are considered to be one of the

most controversial items in the UBC in light of the 1971 San Fernando Valley

earthquake. Many residential structures which had to depend on the principle of

rotation in order to remain standing, fared reasonably well. Generally though,

this type of structure has more than average damage and in some cases collapsed.

Improperly braced cripple stud walls, improperly tied levels of split-level homes

and the principle of rotation are the three principle causes of residential damage

due to earthquakes . The manual will recommend that rotational principles of design be used for one-story detached garages.

5. Sec. 2517 Structural Roof Sheathing : Plywood roof sheathing is used in most

home construction in the U.S., but various other systems are used in the Los Angeles area. Most of these other systems would not qualify as diaphragms. Despite

this, there was little damage to roofs from the San Fernando earthquake. This was probably due to the fact that interior walls stopped at the ceilings and the

roofs only carried their own weight. The manual will, therefore, not specify as

does the UBC that structural roof sheathing shall be designed to support all loads

specified in the Code.

6. Sec. 2518(f) 5 Bracing : All exterior walls and main cross-stud partitions shall

be effectively and thoroughly braced at each end and at least every 25 feet of

length by one of several methods which include let-in l"x4" diagonal braces,

diagonal wood boards, plywood sheathing, fiberboard sheathing and gypsxim board

sheathing panels.

Braced panels shall be 48" in width. Solid sheathing shall be applied continuously on the first story exterior walls of all wood framed buildings three stories

in height. Solid sheathing shall be applied on either face of the first story ex- terior walls of all wood framed buildings two stories in height in Seismic Zone 3.

These braced wall sections shall be located at each end and at least 25% of linear

VI-80
length of wall shall be braced. All vertical joints shall occur over studs and

horizontal joints shall occur over blocking unless panels are 4'x8' applied vertically.

Poor performance of let-in braces was observed after the San Fernando earthquake. Particle-board is allowed by the Code although no shear values are specified. Some of the materials specified for sheathing have no minimiom attachment

specified. The sheathing specified is to be used in Seismic Zone 3 even though

higher shear forces may occur for two-story residential structures in areas having wind loads of 20 psf or higher. Calculations show that if no interior

shear walls are provided, all of the sheathing materials specified for exterior

walls will be overstressed if a two-story building experiences a 20 psf wind load.

If an interior shear wall does exist at the center line of the structure; fiberboard, gypsum board and particle board would still be overstressed.

Calculations also show that the Code requirements for overturning are conservative. Even if no interior shear wall exists and if the exterior walls are of

minimal length, overturning will not be a problem.

7. Sec. 2518(f) 6 Cripple Walls : Foundation cripple walls are framed walls less than

full story height that extend from foundation wall to the floor or roof level.

Such walls shall be braced as required for first-story exterior walls. Solid

blocking may be used to brace cripple walls having a stud height of 14" or less.

Some split-level homes in the San Fernando earthquake were constructed with

cripple walls. Most of these houses suffered severe damage caused by collapse of

the cripple-stud walls. Let-in braces are not adequate and will not be allowed in

the manual.

'8. Sec. 2518(g) 6 Blocking : Rafters more than 8" in depth shall be supported laterally at ends and at each support by solid blocking unless nailed to header, band

or rim joist or to an adjoining stud.

Prefabricated roof truss-rafters are used predominantely in today's housing construction. When roof rafters are used, they generally are less than 8' deep. In

any case, blocking is seldom specified and shear transfer is not achieved. The

manual will probably require solid blocking for truss-rafters and all other rafters.

9. Chapter 37 - Masonry or Concrete Chimneys, Fireplaces and Barbecues : Many differences exist between the Code, HUD-MPS, and other building codes and standards.

The manual will specify requirements as verified by calculations or observations.

10. General Requirements

a. Roof Diaphragms : Roof diaphragms, as opposed to ceiling diaphragms, play a

small part in transferring lateral loads in residential construction. Since the

roof diaphragm is generally more limber than the ceiling, loads in the direction

of framing are taken by the ceiling to interior crosswalls. Specific requirements

for roof diaphragms, in the opinion of the contractor, accomplish little and only

add to cost.

VI-81
b. Roof Framing Blocking : The Code stipulates that blocking for 2x4 through 2x8

roof framing members may be omitted. Normal procedure in California, however, is

to provide such blocking. Blocking for roof trusses is generally not provided in

the Phoenix, Arizona area. Because of relatively low i.hear forces, blocking may

not be needed continuously. Some shear transfer is required and this is difficult

to provide without some blocking. It is suggested that a minimal length (as little

as six feet) of blocking should be provided near ends of each individual roof.

Blocking requirements will be specified for prefabricated roof trusses.

c. Conventional Construction Provisions; These Code provisions imply that all

residential construction can be considered as "conventional." However, many types

of construction are very unconventional and may require further design considerations :

1) Cathedral ceilings with high shear walls. Shear walls may still be four foot

wide and overturning forces may be considerable.

2) Large rooms and long unbraced walls may be critical. Fewer interior walls to

resist earthquakes may be provided.

3) Two-story homes may have insufficient shear walls.

4) Broken roof diaphragms may occur in split-level homes and in other unusual

configurations

.

5) Partial second floors very often do not have adequate shear walls near interior extremities.

Economy of construction must play a very important role when considering modifications to requirements for home construction. Some design requirements in the

Code should therefore be revised in the manual. Other normal engineering procedures not specifically mentioned in the Code will be ignored. Bracing requirements

provided in the manual will obtain the most in struct\aral integrity for the least

cost.

Shear Resisting Materials and Shear Walls ; In order to obtain the most efficiency

from shear resisting materials and to obtain the most liberal interpretation for

the design of shear walls, the following provisions will be contained in the manual.

a. The designer will be instructed to use full dead load for resistence to over- turning moments. Many engineers use a lesser, more conservative value. The Code

specifies full dead load but the Los Angeles City Building Code allows 75% of dead

load. Hold-down anchors have not been required for home construction except in

those isolated cases where engineering design is provided. These anchors frequently cause problems in the field because of misplacement, incorrect installation and the problem of covering bolt heads or nuts, etc. For those reasons,

it is felt desirable to eliminate the need for anchors where at all possible. The

manual will have provisions to allow designers greater latitude in selecting shear

wall lengths, spacing and location to preclude hold-down anchors.

VI-82
b. Shear values for differing materials applied to each side of the same wall are proposed.

c. The height to width ratio for fiberborad, stucco, gypsum lath and plaster and

gypsum board will be increased from 1 1/2:1 to 2:1 as implied in Section 2518(f)

5

of the Code.

12. Sill Bolts : The Code stipulates that sill bolts shall be 1/2-inch diameter em- bedded at least 7" into concrete and spaced not more than six feet apart, with a

minimxam of two bolts per piece and one bolt within 12" of each end. The manual

will require that sill bolts be placed in accordance with shear in the wall which

will increase the number in some cases. Los Angeles City Building Code (1972 Edition)

1. Sec. 91.2305-Horizontal Forces (a) General ; Requirements of this section do not

apply to any conventional framed one-story. Type V building accessory to dwelling,

provided building is not an unusual shape or size and not subjected to unusual

loading conditions. This would exempt most garages. The manual will clarify this

requirement.

2. Sec. 91.2 305(h) Overturning : 75% of dead load may be used to reduce tensile

stresses caused by seismic overturning moments. The manual will specify full dead

load as stated in the review of the UBC.

3. Sec. 91. 2306-Design for Horizontal Force (b) Anchorage :

a. Walls anchored to continuous footings do not need to be anchored to floors

that are within four feet of the footings.

b. When roofs, including their supporting joists, beams or purlins, are con- structed of metal and are not designed as diaphragms to resist horizontal forces

anchors may be spaced at greater then four feet on center. These provisions will

be contained in the Manual.

4. Division 24 - Masonry, Table 24-H, Note No. 5 : Shear walls which resist seismic

forces shall be designed to resist 1.5 times forces as determined by Sec. 91.2305

(d)2. The allowable shears in Table 24-H are lower than the UBC values. The

manual will use shear values from the UBC but will require the 1.5 factor for overturning.

5 . Division 25 - Wood, Table 25-H, -Allowable Shear for Diaphragms and Shear Walls :

This table sets forth allowable shear values for 1" and 2" straight sheathing,

1" and 2" diagonal sheathing, fiberboard wall, sheathing, expanded metal lath

with cement plaster, gypsum lath and plaster, and gypsum wallboard. Several

clarifying notes are appended. Table 25-C contains relative strength values of

bolts, nails, screws and connectors in different species of wood. The shear values

in Table 25-H are at variance with the UBC. However, the format is excellent and

will be incorporated in the manual. No similar table to 25-C is contianed in the

UBC - it therefore will be printed in the manual.

yi-83
6. Sec. 91.2512(b) Design Considerations: Wood diaphragms and wood shear walls shall

be considered flexible and shall not be used to transmit rotational forces unless

otherwise approved. Wood diaphragms in houses are not necessarily flexible, therefore, the manual will present some design criteria.

HUD Minimum Property Standards (MPS) (1973 Edition)

A preliminary review of the MPS has been completed and several sections may be affected

by the manual. In addition some general comments were offered by the contractor which will

not affect the manual directly, but may affect implementation of its requirements. Specific

commentary will not be known until the manual is fully developed. It is intended to then

reference the manual in the MPS manual of Acceptable Practice with applicable commentary.

Conclusion

It has been the intent of the author to present a status report on the research project

undertaken by the Applied Technology Council for HUD. Major pertinent areas of concern and

conflicts between various building codes under which most U.S. housing is constructed, are discussed with an attempt made to present their treatment in the proposed "Manual of Recommended Construction Practice for Earthquake Resistive Residential Dwellings."

The paper was prepared from a review of draft documents submitted by the subcontractor:

Ralph W. Goers and Associates, Sherman Oaks, California to Mr. Roland L. Sharpe, Executive

Director of the Applied Technology Council, Palo Alto, California. This review and analysis

of recommendations are solely the interpretation of the author and do not necessarily reflect the official position of the U.S. Department of Housing and Urban Development, the

Applied Technology Council, or the subcontractor.

VI-84
ATC-.'i-SrAl "HAiiUAL or i;i;com;ii;;;i)i.i) C(>N;;Ti;LK;-noN iio\c;tici; rou

EAKTIICJUAKli WiS 1 ST 1 Vli KtSlDii.NTIAL DUKLLINCS"

ci:N'r.!;AL

1. Wli.Tt building code (or codes) aire used in your area?

2. If tl)c code utilised is Locnl, is it based on:

A. nOCA

B. SoiiChern lluildina Code

C. Uniform liuilding Code

D. OLher (specify)

3. A. Snow load or roof live load is psf.

B. Wind load (not zone) for lowest building is psf.

C. Residential floor live load is psf.

4. Is residence roof and floor framing "designed" (by span tables or

actual engineering)? Yes No

5. Wliat seismic ^onc are you in (if known)? 0 I 2 3

6. Are small commercial buildings in your area required to be engineered

by a licensed engineer? Yes No

If yes, are wind or seismic designs required? Yes No

7. Are residences and tracts plan checked? Yes ___ No

By whom? Licensed engineer

Licensed architect

Unlicensed personnel

8. What is your estimate of the percentage of tract type housing designed

by - Architects X

Building Designers '/. Drafting services

Contractor employee %

Other (specify)

9. What percentage of homes in your area are "split-level"? (Split-level:

two or more floor elevations approximately one-half story apart in

height) 7.

ROOF AND FLOOR FR.\MING

1. Are any systcis other than wood frame used for floor and roof framing?

(Specify) Material

Roof : 7. used

Floor: 7. used

2. Arc roof truss designs normally engineered by the fabricator?

Yes No

3. What Is the weight range of roofing materials used?

Material Woip.ht

Heaviest _________ psf

Lightest . psf

VI-85
/i . Is Moikiiii; licLwLH'ii i.'oiul f r.iiiii iij; lui'inln.' :;; rcqti i ilhI .jI Koof VcE No Floor Yes No

5. Listed in ilcsceiulinf, order of use (nio^t common firsL) wliaC slicoMiinu

ninCcrinl:, ;irc used?

Koof Floor

6. \vl>cit species and grade of lumber are most commonly used for roof and

floor framing?

Roof Floor

Species Grade Species Grade

7. Are you familiar with the terms roof and floor diaphragms? Yes No At what size building do you almost always require they be designed?

8. What percentagcd of roof framing utilizes .prefabricated roof trusses?

7.

WAIX FRAMING

1. Please list the types of vail framing and exterior finishes used in

your area with your estimate of the percentage of tract type homes

built using that system. (Examples: Brick Masonry, Block Masonry

with or without interior furring, v.-ood studs with all walls veneered,

wood studs and fiberboard, etc.)

Materials Pcrcontaso

A.

B.C.

D.

E.

F.G.

2. VJhat types of interior finishes are commonly used in your area? Materials Perccnta'';e A.

B.C.

D.

E.

3. Arc masonry exterior walls in residential construction

A. Fully reinforced

B. Partially rein forced

C. Unrcinforccd

VI-86
Ai'f a)] Mi.i.'.onry walltil lumsi'S (c.stciior ;ind iiiti-rioi) built Willi iiny

/*'<'<|ii(iicy in your orc.i? Yes No

5. Do you require dov/fls to tlio fooLiiit to match vortical rcinforcinu

A. Kcsitlfiicc Yes Ko

n. Commercial lkiildiiii;s Yes No

6. Species and (;rade for v/ood studs arc

7. Arc methods of attaclimont of wood stud wall finish material specified

in your code? Yes No

If yes, do these requirements pertain to residences? Yes No

By your best estimate, are, these requirements met in the field (for

residences)? Yes No

8. What type, size and spacing of sill plate attachment (to foundation

or slab) do you require?

9. Although this question is difficult to estimate, we ask your indulgence. For the average one-story house and the average two-story

house, if a horizontal line drawn across the front and rear elevations (first floor for two-story) such that. as much of the line

as possible . crossed window."?, doors and other openings, what percentage of the line would be across such openings?

One-story Two-story

Rear Elevation

Front Elevation

FOUNDATIONS

1. Are slabs on grade used in your area? Yes No

If so, what is required thickness?

Are they required to be reinforced?

Type of reinforcing?

2. What percentage of homes in your area have basements? '/, Are basement walls - Concrete

Masonry (type)

Other (specify)

Is reinforcing steel required? Yes No What reinforcing? ^

3. What Is minimum size footing required?

One-story Two- story

Depth below grade: _______________

Width:

yi-87
A. U'il.il. 1:; fror.t lino tlc|)lli?

5. Arc ixlcrior fooLini;;; normally

A. Tull widlli pour

1). Koocinp iind CooCinc vmH

C. Oclicr (Specify)

6. Is rcinforcinc norn^cilly required in footinps? Yes No

If so, u'lmt and where

7. Arc houses frequently built in your area with tlie lowest story one

half story bclovi; grade? Yes No

Arc one or more sides frequently at grade when this construction

Is used?

Will you include, please, any "hand-out" sheets or other information

pertaining to residential construction, fireplace construction, foundation

requirements, etc. which your department utilizes?

SPECIAL CONDITIONS IN YOUR AREA — OR OTHER COM>ENTS

*

yi-88
DYNAMIC TESTS OF STRUCTURES FOR OIL TANKS AND NUCLEAR POWER PLANTS

by

Seiichi Inaba

National Research Center for Disaster Prevention

Science and Technology Agency

ABSTRACT

In Japan, it is necessary to build an increasing number of nuclear power plants in

order to overcome the energy problem. Also, an increasing number of large oil tanks have

been built in order to increase the amount of standard crude oil. It has, therefore, become

important in engineering to design aseismic structures for nuclear power plants, oil plants,

gas tanks, and other industrial plants for the purpose of preventing disasters due to the

earthquake. Under these circumstances, several dynamic tests of plant structures have

been conducted by using a large-scale shake table of the National Research Center for Disaster Prevention (VRCDP) in Tsukuba New Town. This report will present in general the

results of dynamic tests on a graphite shielding structure, oil tanks, fuel assemblies of

nuclear reactor, and a container vessel of thin shell.

Key Words: Aseismic Design; Dynamic Tests; Nuclear Reactors; Oil Tanks; Power PlantsShake Table.

VII-1
Introduction

A large-scale shake table, of the NRCDP, was built in 1970. Since then, a number of

vibration tests on structures have been conducted under a joint research program with

engineers of other research organizations. Most of the dynamic test projects have been

related to problems concerned with soils, soil-structures, and soil-structure interaction.

Some of the test results have been reported at a previous Joint Panel on Wind and Seismic

Effects.

Dynamic Test of a Graphite Shielding Structure

The dynamic test of a graphite shielding structure, used for fast reactors, was conducted by the Power Reactor and Nuclear Fuel Development Corporation and the Building

Research Institute in June, 1971, using the large-scale shake table of NRCDP.

The graphite shielding structure is a masonry- type structure, composed of a great

number of graphite blocks to fill the space between the reactor vessel and a safety vessel.

It is necessary to leave gaps between the blocks in order to account for dimensional

changes due to thermal expansion and radiation effects. The graphite blocks are connected

by graphite shear keys, and the outside rows of the blocks are linked by steel pins to the

safety vessel.

Aseismic stability of this structure was to be obtained directly from the vibration

test, because it is difficult to theoretically analyze structures with many gaps. The

half scale model of the graphite shielding structure was built using the same materials as

those of the prototype. Eleven layers of the graphite blocks were piled up and jointed to

a cylindrical cell of 3200 mm in outer diameter, 2270 mm in inner diameter, and 1067 mm

in height.

The model structure was tilted 30 degrees and then 45 degrees on the shake table in

order to apply the lateral force for a static test. Dynamic tests were conducted by

inducing sinusoidal waves of various frequency range including the design earthquake waves. Measurements of stresses in the representative blocks and keys, stresses of the pins,

relative displacements of the blocks, and accelerations of the graphite blocks were made

throughout during the tilting test and the vibration test. After completion of these

tests, an ultimate strength test was conducted by applying static loads, to determine the

ultimate strength of several graphite blocks of 1 or 3 layers and 3 rows connected by the

shear keys and pins.

The results of the ultimate strength test showed that the ultimate limit acceleration

of the blocks corresponds to the collapse mode around the pin hole, (the limit acceleration for the pin) this limit acceleration was 52.4 g and the ultimate limit acceleration of

the blocks, corresponding to the collapse mode around the key way, (the limit acceleration

for the key) was 57.3 g.

The results of the static tilting test indicate that the limit acceleration for the

pin key were 31.8 g and 56.9 g respectively, as obtained from the measured stress values due

to the lateral force of 45° inclination. Therefore, the stress concentration on the model

structure for the actual number of rows was greater than that of three rows. The dynamic

amplification was determined from vibration tests data, and the values for the pin and for

VII-2
key were 6.8 and 3.5 respectively. Thus, the limit acceleration for the pin was 4.7 g

and that for the key was 16.3 g. The vibration test of the 1/2 scale simulation model

shows that the limit acceleration was 4.7 g, which is equivalent to 2.3 g on the prototype structure according to the similarity law.

The results of the response analysis of the reactor building and the safety vessel

supporting the graphite shielding structure indicate that the maximum response acceleration

of a design earthquake (0.15 g at the base of the building foundation) was 0.43 g at the elevation of the graphite shielding structure.

From these results, it is considered, therefore, that the graphite shielding structure was aseismic. The safety vessel had high stiffness and the earthquake response of the

vessel was uniform in the vertical direction. The effect of insufficient number of layers

for the model can be ignored. The design of the graphite shielding structure can be

applied to other areas of greater earthquake design for this reactor, if the pin is reinforced.

Dynamic Test of Fuel Assembly for Nuclear Reactor

The fuel assembly of a nuclear reactor is one of the most important parts of the

reactor, as it contains the boiling water, and, thus, it is necessary to evaluate its dynamic

characteristics. The dynamic test of such a fuel assembly, was conducted by Hitachi and

the NRCDP in June, 1974.

A unit of 4 fuel elements was used for the test, and had a dimension of 4,500 mm in

length, containing 49 fuel rods inside the channel box with dimension of 138 mm and 2.03 mm

in thickness. The model fuel rods were made from lead, which had a density approximately

equivalent to that of uranium dioxide. The four fuel element unit was fixed to the

supporting system inside a steel tank, which had a dimension of 800 mm in diameter and

5200 mm in height.

The lateral vibration forces were applied at two horizontal directions 0 degree and

45 degrees. These angles were between the side of the fuel element and the direction of

the dynamic force. The influence of the inserted control rod insert and the gap between

the fuel elements was studied. The effects of the boiling water inside the reactor core were also examined by filling the tank with water and thus simulating submerged conditions.

The upper joints of the fuel element are composed of channel fasteners with the upper

lattice panel and channel fasteners (designated as a spring joint) was fixed by a spring in

the original design. The tests were conducted using a wedge type fastener in order to

study the influence of the fixed supporting system (designated as a pin joint)

.

The dynamic response of the structure was measured by accelerograms attached to the

fuel assembly, tank, and the shake table. Sinusoidal waves, with a frequency range between

3.0 and 8.0 Hz at a sweeping speed of 0.025 Hz/sec, were induced for the vibration tests.

The input accelerations were between 0.1 g and 0.3 g and were chosen in order to test the

influence of the various vibrational forces on the response structure. The design earthquake response waves, obtained from the analysis, were also applied to the model structure

under several maximum acceleration steps.

VII-3
There was not much difference in the frequency characteristics, for the test results,

when considering high to low frequencies. This means that the non-linearity of the upper

spring joint was not as assumed. The acceleration perpendicular to the direction of the

dynamic force of the vibrating elements was 1/3 of that of the direction of table vibration.

The measured accelerations during the 45 degree vibration test was smaller than those

during the vibration of the 0 degree test, with the difference becoming smaller between

the test values, as the vibration force was increased. The dynamic amplification ratio

was greater for the 0 degree test than for the 45 degree vibration test. For the tests with

the spring joint system, the resonant frequency became higher and the amplification ratio

increased with an associated increase of the applied force. In the case of the pin jointed

system, the amplification ratio of the elements was constant even when the applied input

force was increased. The resonant frequency of the fuel assembly, with inserted control

rods, was not different from that of the structure without control rods. When the fuel

assembly was immersed into the water, the amplification factor of the fuel element with

control rods was reduced to a range between 1/2 to 1/4 of that of the element without control rods. The data indicate that the control rods can cause damping, and thus will reduce

the dynamic response due to earthquake. The amplification ratio obtained from the earthquake vibration tests were between 1.1 and 1.3, and thus the dynamic amplification ratio

of 2.0 used for the actual design of fuel element is conservative.

The results of the dynamic tests show that it is necessary to design the sturctural

fuel assembly, without control rods and with the force direction at 0 degrees. The nonlinear characteristics of the fixed joint system causes no problem on the response analysis

under the comparatively greater design accelerations and also when the fuel assembly is

submerged into the water.

Dynamic Test of Thin Shell Container Vessel

The container vessels for the nuclear reactor are shell type vessels with additional

structure elements inside the wall. It is generally assumed that such vessels will show

a complicated dynamic response due to earthquakes. The dynamic test of a model thin shell

container vessel was therefore conducted by the Hitachi and the NRCDP in July, 1974 after

completion of the fuel assembly act.

The model structure was domed shaped and had a dimension of 2400 mm in diameter, 3600 mm in height and 2 mm thick, and was made of vinyl chloride resin. The model had a circular flange bonded to the bottom of the vessel which was then bolted to the largescale shake table. The model container vessel had supplmentary weight inside the shell

wall and a stiffener belt outside the vessel. The model was subjected to sinusoidal

vibration waves forcing out various frequencies and acceleration.

The accelerations of thin shell in the normal direction and the stresses of the shell

in radial and vertical directions were measured using accelerograms and strain gages.

The results of the dynamic tests indicate that the vibrational behavior of the

this shell type vessel simulates beam type vertical vibrations. The resonant frequency

was approximately 19 Hz, which was not influenced by the supplementary weight or by the

VII-4
stiffener. In the range of the low frequencies shell type characteristic vibrations were noted but the amplifications were smaller than those observed for the beam effect. The

shell response increases the resonant frequencies but reduced the dynamic amplification in

the lower frequency range.

Dynamic Test of Oil Tank

In order to increase the crude oil reserve in Japan, it is necessary to build superscale crude oil storage tanks. The dynamic test of a model oil tank, was therefore conducted in order to confirm the aseismic stabiltiy of floating roof tanks. Such tanks are considered to be suitable for super-scale oil tanks. The dynamic test of a model tank was conducted by Nippon Kokan (NKK) and the NRCDP in February, 1974.

The purpose of the test was to establish the sloshing movement of fluid in the tanks,

examine the stresses in the shell wall, and to design the aseismic floating roof tanks.

The wave heights of the fluid, the dynamic pressure, the velocity, and the strains in the

tank shell were measured to obtain the dynamic characteristics of the tank and the contained

fluid. Sinusoidal, triangular and rectangular waves were applied for the dynamic test, as were earthquake simulated waves of six representative earthquakes.

Three model tanks were built on the shake table. A cylinder type model used for a sloshing test had a diameter of 2860 mm, a height of 1219 mm and a steel plate thickness

of 4.5 mm. The rectangular type model also used for sloshing test had a length of 2000

mm, a width of 1000 mm, a height of 1000 mm, and a steel plate thickness of 4.5 mm. The

third model was a cylinder type which had a diameter of 2190 mm, a height of 914 mm, a

steel sheet thickness of 0.4 mm and was constructed for the dynamic tests. The cylinder

type model tank used for the sloshing test was also used to determine the damping effects

of a floating roof with damping fins. The results of the dynamic tests showed that the

stresses in the cylinder wall due to the vibrational force were greater than those

assumed from the analyses. The wall stresses due to sloshing fluid however were less

than the stresses assximed from the analysis. The natural frequency of the sloshing fluid

was in good agreement with those obtained from the theoretical analysis . When sinusoidal

waves representing the natural frequency of 0.5 Hz for the first mode, 1 Hz of the

second mode, and 1.2 Hz of the third mode were applied to the model tank, the sloshing

phenomena in resonance were caused by the vibration of the sinusoidal waves. No sloshing

phenomenon was observed in the range of frequency which were different from the natural

frequencies

.

In order to reduce the sloshing movement of the fluid surface, dynamic tests were

conducted for the following cases; (1) using a floating roof, with fins for damping, (3)

using a floating with water on the top of the roof. The result of the tests with a floating

roof, showed that the damping coefficient was not different from that of the free surface

fluid. In the case of the test for a tank with a floating roof with damping fins, the

damping coefficient of the sloshing was three times as much as that of free surface sloshing.

In the case of the test for tanks with a floating roof with water on the top as weight, the

damping coefficient was two times as much as that of the free surface sloshing. This means

VII-5
that the damping fins are quite effective in reducing the sloshing movement.

Conclusion

This paper has briefly described the dynamic tests using the large-scale shake table

of the NRCDP, and the obtained on structures for nuclear power plants and oil tanks, these

tests are quite important in order to prevent disasters due to the earthquakes.

The oil tank fires, due to the Niigata Earthquake in 1964, and the crude oil leakage

and pollution caused by the collapse of tanks in to the Inland Sea in Okayama in 1974, and

the possible Kawasaki Earthquake predicted from data on ground upheaval , have aroused

concern in the field of earthquake engineering and in the aseismic stability of structures

in industrual plants. However the industry and private organizations are primarily

responsible for the design of such structures and the government institutions can only

regulate the design of such structures. It is therefore necessary to develop the theoretical analysis of the dynamic behavior of plant structures and also to test large-scale

structure on a shake table. In addition, it is necessary to improve the shake table,

development of techniques for dynamic tests, and studies regarding simulation and

modeling.

The Resources and Energy Agency of the Ministry of International Trade and Industry

has appropriated part of the funds to build a new large-scale shake table in their budget

for the fiscal year of 1975-1976. The shake table, of dimension of 30 m square can support a test structure of 1000 tons and was built for the dynamic testing of structures

for future nuclear power plants. However, there are still many problems concerned with

the construction of such super-scale shake tables.

It is hoped that by exchanging technical information through the Joint Panel on

Wind and Seismic Effects, many of these problems can be solved.

Bibliography

Kawaguchi, 0., Watabe, M. , et al., "Special Aseismic Design Engineering and Development

for Fast Reactor", Proceedings Vol. I, Engineering of Fast Reactors for Safe and Reliable

Operation, Karlsruhe, 1973.

VII-6
SHEET PILE FOUNDATION AND ITS STRUCTURAL

CHARACTERISTICS AGAINST HORIZONTAL LOADS

by

Kenji Kawakami

Tadyoshi Okubo

Keiichi Komada

and

Michio Okahara

Structures & Bridges Division

Pxablic Works Research Institute

Ministry of Construction

ABSTRACT

The general response of sheet pile foundations, relative to analytical and experimental

studies, are discussed. Test results are given, in addition to the development of general

design equations.

Key Words: Foundation; Lateral Loads; Sheet Piles; Structural Analysis; Tests.

VIT-7
1. Introduction

Sheet pile foundations consist of steel piles with joints, driven into the ground to

form a closed unit. This is a unique foundation structure and was developed by the Japanese

steel manufacturers. Since this foundation uses steel piles, it has advantages in that the

work can be mechanized and completed within a shorter period of time and that the dimensions

and penetration length of the foundation can be freely chosen. The enclosure of the sheet

piles also can be used for a cofferdam by making their joints watertight.

The sheet pile foundation was first used in 1965, for the foundation of blast furnaces,

and was used for bridge construction in 1959. Owing to its outstanding advantages, the

sheet piling method has been used in the construction of some 148 foundations for 40

bridges, and it will probably be used more frequently in the future. This method has primarily been used for such cases that cannot use cassions or piled foundations, because the

ground and water depth conditions are poor and the dimensions are large.

The sheet pile foundation is considered to be a structure having mechanical characteristics intermediate between the caisson foundation and piled foundation. It is therefore,

necessary to establish more efficient design of the sheet pile foundation, by making

further investigations on its mechanical characteristics.

In this paper, we would like to discuss the structure of the sheet pile foundation,

the presently used design method'''^ , past performances, and the results of in-site experiments. Also to be presented will be the results of an analysis we have made in an attempt

to clarify the structural characteristics of the sheet pile foundation. The sheet pile

foundation has such complicated structural characteristics , that we have advanced only one

step in its investigation.

We will not discuss the use of the sheet pile foundation for cofferdam, in this paper.

2. Sheet Pile Foundation Structure

2-1 General Definition

The sheet pile foundation is defined as a structure that forms a circular,

elliptical or rectangular enclosure with steel-pipe or H-shaped steel piles driven

into the ground. Their heads are rigidly connected, and the joints are so treated

that the required horizontal resistance and vertical bearing capacity is developable.

As will be seen from the examples given later, it can be said that steel-pipe piles

can be used in many instances. As for the treatment of the joints, they are grouted

with mortar to prevent the passage of water, because the sheet pile foundation is also

used to serve as a cofferdam. The grouting of the joints with mortar, although it is

a technically difficult work, is employed because the weak points of the sheet pile

foundation lies in the joints and its strength can be increased by the treatment of

such joints. Much is expected in the development of efficient joints for the future.

2-2 Material

The material that is used for the sheet pile foundation is steel, which has

sufficient strength, quality and sectional characteristics to satisfy its structural

configuration. As a rule, the steel shall be over 8 mm in thickness. The sheet pile

VII-8
foundation utilizes steel piles or H-shaped steel sheets, whose shapes and dimensions

are described at the end of this paper. The shapes of steel piles and joints are shown in Fig. 2-1 and their dimensions and weights are given in Table 2-1. Typical

values are given in the figure 2-1. The steel used in the making of the sheet piles

are defined by JIS A5528 and is as follows.

Steel-pipe piles: SY24

H-shaped steel sheet piles: SY30, SY40

The allowable stress intensities for the steel piling, as given in the chapter devoted to surveying and designing in the Specifications for Substructure Design of

Highway Bridges, are given herein in Table 2-3.

2-3 Structural Types

The sheet pile foundations can be classified by shape into two types, that is,

the well type which is formed with the piles of equal length driven to the bearing

stratum and the pronged type in which some of the piles are driven to the bearing

stratum. These two basic forms of sheet pile foundations, can be further divided

into several types according to the cross sectional contour, number of foundations,

and joint grouting method as shown in Table 2-4.

As was mentioned previously for the well type, all the piles are driven to the

bearing stratum and the head of piles are rigidly connected by a footing. However,

when the bearing stratum is located at a relatively great depth, it has greater vertical load bearing capacity than horizontal capacity. In such cases it is more economical to employ the pronged type, in which some piles are driven to the bearing stratum

while others are stopped midway between the ground surface and the bearing stratum,

thereby reducing the quantity of steel piling.

The sheet pile foundations can have such cross sectional contours as circular,

oval and rectangular, as illustrated in Fi.g 2-2. In the case of caisson foundations,

the cross sectional contour greatly influences the speed at which the construction

work can be conducted. In this method, however, the work differs little from the

conventional piling work, therefore the cross sectional contour arrangement has little

influence on the speed of construction.

The sheet pile foundations can be divided into five types, as illustrated in

Fig. 2-3, according to the differences in the head and coffering. Class (1) is

called the "rising" type, in which the footing is above the water surface. Class (2)

is called the "closing" type in which a closing wall is built independent of the

foundation body and footing. Class (3) is called the "coffering" type, in which the

sheet piles rise above the water surface to form a cofferdam and the coffering parts

of the sheet piles are above the water, after the foundation body and footings are

completed. Class (4) is called the "semi-rigid" type, in which the inside of the wall

is partially excavated and filled with concrete. Class (5) is called the "rigid" type,

in which the inside of the well is excavated down to the bearing stratum and filled

with concrete.

VII-9
2-4 The Development and Characteristic Features of Sheet Pile Foundations

The sheet pile foundation was used as a foundation for the first time in conjunction with the blast furnaces at the Mizushima Iron Works of Kawasaki Steel Corporation.

In addition, after repeated tests on models and actual sheet pile foundations, this

method was used in the construction of bridges for the first time on the Ishikari Kako

Bridge (Hokkaido), 1969) and then for the Omigawa Bridge (Chiba Prefecture, 1970) and

the Senbonmatsu Bridge (Osaka City, 1970)

.

As noted the first application of this method was in the construction of a blast

furnace foundation, with the development and research of the sheet pile foundation con- ducted by several major steel manufactures. Apart from the blast furnace foundations,

this method was also used dn various civil engineering works, such as dolphins and

scale pits at the iron works. At that stage, development and research projects were

conducted, thereby developing the fundamental techniques for its application in the

construction of bridge foundations.

The sheet pile foundation, is considered to have a rigidity close to that of

the caisson foundation, and can be built almost in the same manner as the piled

foundation and is well suited for labor-saving and rapid construction work. This

method has so many advantages, that the Ministry of Construction granted research subsidies to the steel manufactures in 1969, in order to encourage their research relative

to the workability and structural characteristics of sheet pile foundations. In this

relation, the steel manufacturers took the initiative to form a sheet pile foundation

research committee which included university professors and government officials, in

(3) order to conduct the four research projects . These projects were; (1) investigations on pile driving accuracy, (2) the water seepage prevention, (3) the integrity of

piling and concrete grouting, and (4) vibration characteristics. As a result of these

studies, the "Specification for Designing and Construction of Sheet Pile Foundations"

was published in January, 1972. Although this is a Specification, which was developed

from nongovernment basis unlike the technical standards set up by the Ministry of Construction i.e. the Specification for Highway Bridge Design and the Specification for

Substructure Design of Highway Bridges, it has been widely used in the actual design

and building of bridge foundations.

One of the most outstanding features of the sheet pile foundation is that the

coffering method can be used. As a matter of fact, this particular type has been

used in most of the Sheet pile foundations constructed. Fig. 2-4, shows the steps involved in construction of this type of the sheet pile foxindation. The building of a pile foundation in water requires a cofferdam, a great deal of money, and takes a long

time to complete the work when the water is deep. The mechanical stability of a cofferdam is not fully understood. In this foundation type, the cofferdam is built with the

same structural material as the foundation body, therefore, it has a very high degree

of safety. It is possible to further increase the safety by filling the gaps between

the cross beams of H-shaped steel sheet piles and steel-pipe piles, thereby integrating

the timbering and cofferdam. If the sheet pile foundation is to be effective as a

VII-10
cofferdam, it is necessary to have watertightness of the joints, and have a proper

technique in cutting of the steel pipe piles in water in order to have an efficient

and economical product.

3. Designing of Sheet Pile Foundations

3-1 Fundamentals of Design

The basic principles to be considered in the design of sheet pile foundations

are listed below;

(1) The maximum ground reaction on the well bottom and at the distant ends of the

prongs of sheet pile foundations must not exceed the allowable bearing capacity

of the ground at the given position.

(2) The shear resistance on the well bottom of a sheet pile foundation must not

exceed the allowable shear resistance between the well bottom and the ground.

(3) The displacement of a sheet pile foundation must be considered, by considering

the allowable amount of displacement that is determined in relation to the

superstructure

.

(4) The intensity of stress, in the various parts of a sheet pile foundation,

must not exceed the allowable stress intensity.

(5) Since the horizontal resistance of a sheet pile foundation is treated as a beam on an elastic foundation, as in the case of a piled foundation, there is no need to consider the ground reaction on the front surface of the foundation.

(6) In the case of a sheet pile foundation, whose displacement is large and

natural frequency is long, it is preferable, if the responsible engineer deems

it necessary, to make a dynamic analysis, such as analysis will take the superstructure into consideration, and to make a comprehensive study of displacements

and stresses in the structural members.

3-2 Determination of Load

The loads to be used in the design of sheet pile foundations shall comply, as a

rule, with Chapter 2 "Loads" in the part devoted to investigations and design in general

in the "Specification for Substructure Design of Highway Bridges" and Chapter 3 "Loads"

and Design Conditions to be Considered in Earthquake Resistant Design in the "Specification for Earthquake Resistant Design of Highway Bridges".

The loads to be considered include live loads, dead loads, earth pressures, water

pressures, buoyancy, effects of temperature changes, effects of earthquakes, snow load

etc.

3-3 Allowable Vertical Bearing Capacity

The allowable vertical bearing capacity shall be determined, taking into consideration the degree of importance of the structure, as follows;

(1) Vertical load test

(2) The formula for statical bearing capacity

(3) Estimated values when the values are proven correct

(4) Formula for the statical bearing capacity

VII-11
When calculating the allowable vertical bearing capacity, several formulas,

not merely one formula, should be used so that a comprehensive study may be made from

the results of all such calculations. The allowable bearing capacity is a value to be

obtained by dividing the ultimate bearing capacity, obtained from the statical formulas,

by a safety factors given in Table 3-1.

The conception of the bearing area, when calculating the bearing capacity of the

ground and the bearing capacity calculation formulas, are shown in Table 3-2. The

symbols used in Table 3-2 are as follows;

Ra: Allowable bearing capacity of a Aheet pile (kg)

Safety factor

Length of penetration of well (cm)

Length of prong (cm)

External circumference (cm)

Circumferential length of prong (cm)

Number of sheet piles forming the well

Number of prongs

S

£

1 £

2U

u

o

n

1 n

2 qd Ultimate bearing capacity of the ground at the distant end of sheet pile

(kg/cm2)

2 f: Frictional force on the circumferential surface of well (kg/cm ) 2 f : Frictional force on the circumferential surface of prong (kg/cm ) o 3-4 Horizontal Resistance

(i) Flexural rigidity

The flexural rigidity is considered to vary according to the treatment of the

joints between the sheet piles, the treatment of the heads of sheet piles, the

type of soil, and the shapes and dimensions of the sheet pile foundations.

As a matter of convenience, the flexural rigidity can be obtained by using

the following formula as obtained from the experiments;

"l "1

E = E( Zli + yZAi y^) (3-1)

i=l i=l ^

where

6 2 E: Elastic modulus of sheet pile (2.1 x 10 kg/cm )4 I: Moment of inertia of sheet pile foundation (cm ) 4

Ii: Moment of inertia of i- th sheet pile (cm )

n^: Number of sheet piles

2 A^: Sectional area of i-th sheet pile (cm )

yi: Distance of i-th sheet pile from the center axis of sheet pile foundation

(cm)

v: Composite efficiency of sheet pile foundation (0 to 1.0).

The reference values are shown in Table 3-3.

VII-12
(ii) Coefficient of ground reaction

The coefficient of the ground reaction, is defined as the split line

gradient of the load-displacement curve. That is,

K = £ (3-2)

where

3 K: Coefficient of ground reaction (kg/cm ) The ground reaction coefficients include horizontal ground reaction

coefficient (K^^) , vertical ground reaction coefficient (K^) , and horizontal shear spring coefficient (Ks)

.

2 P: Load intensity (kg/cm )

y: Amount of displacement (cm)

(1) Method of estimating the ground reaction coefficient, in the well type of

sheet pile foundation

o Method of estimating the horizontal ground reaction coefficient

^1 ~ 'SlO^Io^ ^ =12.8 Kj^qB~ ~ (3-3)

where

3

K : Horizontal ground reaction coefficient (kg/cm ) HI

^H0 K„„:= This is the horizontal ground reaction coefficient (kg/cm"^) which is

equivalent to the value obtained by the plate loading test using 30 cm diameter rigid disk and can be expressed by the following formula including the 20% share by the side.

1 2

K,,^ = ^rr- aE = 0.040 aE

HO 30 o o

B: Load width (cm)

3 E : Modulus of deformation of groxind 9kg/cm ) measured or estimated by the

o methods shown in Table 3-4.

a: Coefficient which are used normally or at the time of earthquakes.

Shown in Table 3-4.

Vertical ground reaction coefficient

- ^

4 Rv — —

K , K ^ = K (^) = 12. 8K B 3 (3-4)

vl, v2 vo 30 vo V

where

3 K , : Coefficient of vertical ground reaction at the well bottom (kg/cm ) vl

3 K ^: Coefficient of vertical ground reaction at the prong end (kg/cm ) v2

VII-13
K : The vertical ground reaction coefficient (kg/cm ) equivalent to the vo value obtained by the plate loading test using a 30 cm diameter disk and

can be obtained by the following formula.

K = — a E 0.033 a E vo 30 o o

B^: Equivalent load width of foundation (cm) , which can be obtained by the

following formula.

Bv = Av

E : Modulus of deformation of groxind measured or estimated by the methods

° 3 shown in Table 3-4 (kg/cm )

a: Coefficients to be used normally or at the time of earthquakes. Shown

in Table 3-4.

A : Base area surrounded by the outer and inner circumferences of the

2 foundation (cm ) When the ground beneath the base of the sheet pile foundation changes in the

direction of the depth, or when there exists a particularly weak stratum, the following

should be considered;

o Horizontal shear spring coefficient

I K = A K 1

s v

3

K^: Shear spring coefficient of the well bottom (kg/cm )

(iii) Calculation of horizontal resistance

Assuming the sheet pile foundation to be a beam on an elastic foundation, its

horizontal resistance shall be obtained by the following formula.

EI^ = -ky (3-5)

dx

where

3 k: Horizontal ground reaction coefficient (kg/cm )

y: Amount of displacement (cm)

x: Depth from the ground surface (cm)

Both the well type and the pronged type of the sheet pile foundation shall be

such a structure that there exist a horizontal spring, vertical spring and shear spring

on its front, side and bottom. Furthermore, in the case of the well type, the bottom

shall receive a vertical ground reaction proportional to the vertical displacement and

a shear resistance proportional to the horizontal displacement.

VII-14
When the penetration length is short for its rigidity, and the shear deformation

is so large in amount that it cannot be ignored, the foundation must be designed by

taking the shear deformation into consideration.

(1) Design of the well type sheet pile foundation

A mechanical model of this type foundation, is illustrated in Fig. 3-1.

o Basic Formulas

The following, are the basic equations relative to the model shown in Fig. 3-1.

EI

^ = -P

d X

where

P = y^ = Dy^ (cm)

D: Load width of sheet pile foundation (cm)

y^, y^: Displacement (cm)

The following are the general solutions to equations (3-6)

For the part above the ground

3 2 y = A X + B X + C x + D

o o o o o

For the part under the ground

-6x

y^ = e (A ^ sin 3x + cos 3x)

6x

e (C^ sin 3x + D^ cos 3x)

(3-6)

(3-7)

The unknown constants can be determined by substituting the following

boundary consitions into the general equations;

VII-15
X =~h : -Elyo" = -Mo

-Elyo'" = -Ho

X = 0 : yo = yi, yo' = yi'

II 11 ii|

X- ^ :-EIyi"=K^iIiyi'

-Elyi"' =-K3 AVYi

However, when a tensile force is acting on the well bottom, the solution must

consider (moment of inertia well bottom) , and A^' (effective loading area of well

bottom) , which ignore the part where the tensile force is acting.

3-5 Section Forces

The stress intensity in the vertical direction of the well type sheet pile

foundation can be calculated, using the section forces of the well. Using the

vertical force the horizontal force and bending moment working on the top of the

well is then obtained, and then the stresses. When the composite efficiency exceeds

0.5, the equation (3-9) shall be used and when it does not exceed 0.5, equation

(3-10) shall be used.

> (3-8)

a =

a =

Vq

nl Ao

Vo

ni Ao -

M,

+

V

(3-9)

(3-10)

1=1

where

2 O: Stress intensity in the vertical direction (kg/cm ) V : Vertical force working on the top of the well (kg) ° 2 A^: Sectional area of a sheet pile (cm ) M : Moment working on the top of the well (kg/cm) °

4

I: Moment of inertia (cm ) with considering the composite efficiency (y)

y: Distance (cm) from the central axis of the well to the point where stress is

calculated

n^: Number of sheet piles in the well

Z^: Section modulus of each sheet pile (cm"^)

Z.: Section modulus of the foundation when y = 1.0

VII-16
n: Bending moment distribution constant which is determined by the composite

efficiency, as shown in Table 3-5, for example.

(4) Table 3-5

4. Past Pile Foundation Construction Projects 0. 5 0. 93

Important examples of sheet pile foundations, constructed prior to January 1975,

are shown in Tables 4-1, 4-2, and 4-3. These tables show the project name, purchaser,

location of work, starting date scale of work (basic sectional dimensions, sheet pile dimentions, niomber of foundations), the type of sheet pile (type, type of joints, length of

joints) , and the basic type. Table 4-1 shows those cases in which steel-pipe piles were

used in the construction of bridge foundations. There are 37 bridges and 126 sheet pile

foundations. Table 4-2 shows the cases in which steel-pipe piles were used in the con- struction of blast furnace foundations and other foundations. There are 14 cases and 59

foundations. Table 4-3 shows the cases in which H-shaped steel sheet piles were used.

There are 6 cases and 32 foundations. A total of 148 sheet pile foundations were con- structed for 40 bridges. As seen from these cases of sheet pile foundation projects,

steel-pipe piles were used for most cases. There were very few cases in which H-shaped

steel piles were used, and almost none in recent years. It is also noticeable that there

is an increasing number of the cofferdam type, which is an important characteristic

feature and merit of the sheet pile foundation.

Figs. 4-1 through 4-4 show the histograms, obtained from the data given in these

tables. Fig. 4-1 shows a histogram representing rectangular or oval sheet pile foundations,

using their largest width. Of the bridge foundations built with steel-pipe piles, there are

16 such fovindations ranging from 10 to 15 m in width. Fig. 4-2 shows the circular-shaped

sheet pile foundations, arranged by their diameters. In this instance, these are the only

cases in which steel pipe piles are used. The foundations of this type, range from 5 to 10 m in diameter, and are used most often for bridges. There are as many as 30 such cases.

Fig. 4-3 shows a histogram arranged by the sheet pile length. The sheet piles used most

often for bridge foundations are 25 to 35 m in length. Fig. 4-4 shows a histogram

arranged by the steel pipe diameter. The largest in number are the steel-pipe piles ranging

from 800 to near 1000 mm in diameter. As many as 57 foundations fall into this category.

5. In-site Tests of Sheet Pile Foundations

Tables 4-1 through 4-3 show the sheet pile foundations which were constructed in the

past and in-site tests (loading tests, vibration tests, in-site measurements of stresses),

(4) . where the test results are rearranged in Table 5-1 , including 11 cases.

We will now discuss the results of tests conducted on the Omigawa Bridge , in which

extensive data were collected. Fig. 5-1 shows a cross section of the foundation. The

pile length varies from 42.6 m to 24.5 m and the foundation is of the pronged type. Fig.

5-2 shows the column section of the ground. Fi.g 5-3 shows the load-displacement curve.

The load was imposed in a horizontal direction around the middle of a bridge pier. The

displacement measured was located in the horizontal direction at the top end of the bridge

pier. When the load was 200 tons, the displacement was about 13 mm. It is noted that the

curve is essentially straight, and it can be said that the test was conducted within the

VII-17
elastic limits. Fig. 5-4 shows the load-angle of slope curve. The angle of the slope was measured at three points, that is, the top of the bridge pier, the top end of the top plate,

and the bottom of the top plate. Fig. 5-5 shows the distribution of strains in the bridge

pier. Fig. 5-6 shows the distribution of stresses in the sheet pile well. Fig. 5-7 shows

the horizontal response curve for the top of the bridge pier. The natural frequency during

this time was about 6Hz. Fig. 5-7 shows the vertical response curve for the bridge pier.

The natural frequency at that time was about 4.5Hz.

6. Joint Characteristics of Sheet Pile Foundations

6-1 Purpose of Joint Strength Tests

The sheet pile foundations consist mainly of steel-pipe piles driven into the

ground to form a circular, oval or rectangular enclosures, with the piles being connected with one another by means of joints. Because of this connection, their structural

characteristics are greatly influenced by the joints. This may be understood from the

fact that the use of even the joints which have very low rigidity, gives the foundation a rigidity far greater than when no joints are used. However, the rigidity of

the foundation as a whole is naturally smaller than when the joints are the same as in

the case of the main pipe piles. The analysis of such a system taking into account

the joint characteristics, is obviously an important item to be considered in designing

sheet pile foundations. As was discussed in Chapter 3, the moment of inertia of the

entire foundation is evaluated by the use of the "composite efficiency", and then

assuming the foundation to be a beam on an elastic foundation. This method is very

advantageous, due to its simplicity when the length is sufficiently large in comparison

to the diameter. However, as the ratio of diameter to length approaches 1, it is

difficult to apply this method of solution. This is because the shear deformation of the

joint is ignorned.

In order to consider the joint characteristics in the design of a sheet pile

foundation, it is necessary to know the shear strength characteristics of the joint.

There are various tests that can be performed to evaluate the shear strength of the

joint. Two such tests have been conducted, as will now be described. In one test,

three piles were connected in parallel and a punching load was applied to the middle

pile in order to make a direct measurement of the shear strength of the joint. In

another test, two piles were combined, one on top of the other to form a beam and a

load is then concentrated in the middle thereof to determine the shear characteristics

of the joint.

6-2 Direct Shear Test

As shown in Fig. 6-1, three piles were connected in parallel, with their joints

grouted with mortar and a punching load applied to the middle pile. This test was

conducted in order to obtain the relationship between the load and the amount of

displacement of the piles in relation to each other. The shear stress in the joint

was calculated, assuming a fracture between the steel material of the joint and mortar.

There are many different shapes of joints. Used in these tests were pipe type and the

CT type joints, which are shown in Fig. 6-2. The fractures considered are as illusVII-18
trated, assuming the fracture of the joints to be the minimum area of contact, between

the joint material and mortar. (The bond strength of mortar is sufficiently small

compared with the compressive strength)

.

The results of the tests conducted on the pipe type joint will now be discussedRelative to the fracture of the pipe type joint, the shear strength per unit

length Q and the shear stress per unit area can be easily calculated by the following

formulas

.

Q (kg/cm) = ^

2 P

T (kg/cm ) =

(6-1)

2TTDL

where

:

D: Joint pipe diameter (cm)

L: Joint pipe length (cm

The results of the direct shear test will be graphically shown as the relationship between the superimposed load P, shear strength per unit length Q, shear stress

per \anit area T and the amount of displacement between the joints. The amount of displacement and deformation between the joint and mortar, where they are in contact with

each other, increased almost linearly until the load reached about 25 tons. Thereafter,

it increased sharply and ruptured when the load was increased to about 50 tons . This

occurred when maximum shear strength Q and maximum shear stress T were about 250 kg/cm

2 and 5 kg/cm , respectively. The compressive strength of the mortar which was used to

2 grout the joints was a = 180 kg/cm . 5 2 Assuming that the mortar's Young's modulus is E = 2.1 x 10 kg/cm , the modulus

of rigidity of the standard mortar G therefore is as follows;

E 2.1x10^ „ „ ,^5 , , 2 ,^ ON G = = :j = 0.9 X 10 kg/cm (6-2) ^ ' 2(1+^)

6

However, from the test results, the modulus of rigidity G' can be calculated

using the shear stress per unit area T and the amount of displacement of the joints

G' = T/Y = = 2.47 X lo"* kg/cm^ = (6-3)

1x10-^24.7 ^'^^

From this result, it can be stated that the modulus of rigidity of the joints

is considerably lower than the modulus of rigidity of the mortar.

VII-19
6.3 Bending Shear Test

As illustrated in Fig. 6-4, two piles were combined, one on top of the other, with

the joint grouted with mortar to form a beam. A concentrated load was then applied in the

middle of the beam, which was freely supported in the bending shear test. Ten such tests were conducted. Various shapes of joints were used in the test, but consisted primarily

of the pipe type joint and the CT type joint, which were also used in the direct shear test.

The results of only one of the tests will be given herein due to space limitation. The

relationship between the shear stress in the joint and the rate of decrease in the shear

stress for the condition of a rigid connection is obtained from the test results. Therefore, a method which gives such a relationship obtained from the load-strain curve and a method of obtaining the information from the load-strain distribution in relation to the

shear strength transmission coefficient a will be presented.

(1) A Method of Obtaining the relationship from the Load-strain Curve

The shear strength transmission coefficient a is defined as follows:

r(kg/cm2) = a r' (0 < a ^

1) (6-4)

in which T is the joint shear stress in which the rigid connections are imperfect,

and t' is the joint shear stress between the two piles with rigidly connected- The

moment of inertia I would be as follows;

o

where

;

lo = 21- + 2A'(^)2 = A'd^ (6-5)

2 A' : Sectional area of a sheet pile (cm ) 4

I': Moment of inertia of a sheet pile (cm ) When a concentrated load P is imposed in the middle of the beam, the deflection

of the centerpoint 6 can be obtained as follows;

i) a = 0 (No joint connection)

96 EP

ii) a = 1 (Perfect rigid joint connection

P 3 ^ 48 EIq

iii) 0 < a < 1 (Imperfect rigid joint connection)

'> = a P A' 2

2 lo

V )

VII-20
From the equilibrium of moment

dM

da; (6-8)

where

:

At this time, the deflection of the center point is

8 = • ft (6-9)

48 EI'

The joint shear strength Q is

(6-10)

Thus, we can obtain the relations of P, 6 and a.

(2) A Method of obtaining the Relationship from the Load-strain Distribution

Assuming two cross sectional strain distributions of the composite beam, that is,

when the joint is imperfectly connected and when the joint is perfectly connected, the

joint shear strength Q and the shear transmission coefficient a can be obtained as

follows;

6 Strain in the case of

imperfect joint connection

Strain in the case of perfect rigid

joint connection

Fig. 6-5 Strain Distribution

(6-11)

a is determined from the strain distribution is defined as follows

VII-21
a =-

0 e

e d X

' d X

(6-12)

As shown above, the shear strength transmission coefficient a can be obtianed by two

different methods from test results.

The bending test is illustrated in Fig. 6-6. In this test, the joint was a

pipe type, and the relative size of the joint for the main pile is larger in comparison

to an actual sheet pile foundation. The test results are given in Figs. 6-7 through

6-11. Fig. 6-7 shows the load-central deflection curve. It is seen that the experimental curve is approximately midway between the calculated values, in the case of

rigid joint connection, and of no joint connection. When the steel material begins to

2 yield, (the stress is about 2400 kg/cm ) , the deflection increases substantially com- pared to an increase in load. Fig. 6-8 shows the load-deflection in the axial direction of the pipe. The values are measured to the left and right of the supports.

Fig. 6-9 shows the load-strain distribution at a point 1 m away from the center point.

When the load is small, the strain distribution is such that plane bearing occurs but

as the load increases, the upper and lower piles gradually begin to show an independent

strain distribution.

The relationship between the load and the shear strength transmission coefficient

is shown in Fig. 6-10. The solid line, in this figure represent what was obtained

from the load-strain curve given in Fig. 6-7. The broken line represents what was obtained from the load-strain curve given in Fig. 6-9. These two lines should agree

with each other, but actually are wide apart from each other.

Apart from the above-mentioned test, ten other types of tests were conducted,

and will now be described.

When the joint was not grouted with mortar, the strain distribution showed that

the upper and lower piles behaved as an almost perfect single body. Examination of the

load-strain curve shows that the experimental values are roughly in agreement with the

calculated values for the case when there is no rigid joint connection.

None of the tests showed the results in which the two load-shear strength transmission coefficient curves were nearly in agreement with each other, but the two curves were considerably wide apart from each other. There seems to be a tendency that the

shear strength transmission coefficient, which was obtained from the load-strain curve,

is smaller than the actual value. The average of the load-shear strength transmission

coefficient curves, obtained by the tests, is represented by the curve shown in Fig.

6-11. To obtain this curve, the shear strength of the joint (the shear strength

when the joint is rigidly connected) was obtained assuming a joint fracture and it was replaced by the load.

VII-22
7. New Approaches to Structural Analysis for Horizontal Resistance of the Sheet Pile

Foundation

7-1 Linear Interpolation Method

(1) Method of Analysis

The method of analysis to be used is based on the curve given in Fig. 6-11 which

was obtained in the preceding chapter. The analysis of the sheet pile foundation with

rigidly connected joints is obtained by U.L. Chang's formula. This formula is obtained

by the group of piles method when the joints are not connected. They are linearly

distributed by the shear strength transmission coefficient and thus the obtained values

are taken as the analytical solution of the sheet pile foundation.

This value of a is determined from Fig. 5-11 by obtaining the joint shear stress

when the joint is assumed to be rigidly connected.

The section forces of the sheet pile foundation when horizontally loaded can be

obtained as follows.

(i) Section forces when the joints are rigidly connected

^= 4EI (7-1)

where:

Moment of inertia of the entire sheet pile (cm ) 4

Moment of inertia of an individual pile (cm ) 2 Sectional area of an individual pile (cm ) Distance from the centroid of the whole piles to the centroid of the i-th

pile (cm)

The section forces of the whole piles at an arbitrary point is

B.

M^(z)=- exp z) {Ai(„ (cos/?, 2 + sin/?, 2) +— sin/?, 2}

^1

S,(2)= «xp (-/9^z){H^ (cosA,2 -sin/S^z ) - 2A,^oS'"/*i^ )

(7-2)

The bending moment, axial force and shear strength of an individual pile at an

arbitrary point are

VII-23
N ^-y AM,(z)

1 1 I' '

'

1 + K,

where

=,f { (-1 -y } 2

(7-3)

(ii) Section Forces When the Joints Are Not Connected

Analysis shall be made by the use of the group pile displacement method,

analysis can be made by determining the footing displacement

.

If

3

?

2

The

K,

4

(7-4)

we obtain

0

0 0

0

d

y i'e J

(7-5)

Hence we can determine the section forces of an individual pile at an arbitrary

point from the value of

jM^^ =_ ,xp (-^^O {PM^^ (cos/3 + sin/9^z) +^^-^ sinfi^z

N =PN 2 I I 0

PB

(7-6)

VII-24
where

:

(7-7)

(iii) Section Forces When the Joints are Imperfectly Connected

The section forces can be obtained by the use of a, and thus linearly

distribute the section forces of an individual pile at an arbitrary point which

have been determined by (i) and (ii)

.

N^aN + ( 1 - d) /V ^ =°=^. + (1-0!) S^^

(7-8)

In this case, where the linear interpolation method is used, the shear stress of

the joint is needed for determining a and it is necessary to take a representative

value at the section. In determining the representative value, a value of 2/3 of

the maximum shear stress is assumed, and the sectional shear flow is distributed in a

parabolic form.

(2) Comparison of the Measured Values and the Analytical values of Stress Intensities

in the Body of a Sheet Pile Foundation

A comparative study of the measured and analytical stress intensities in the

sheet pile foiindation was made. The analytical values were obtained by use of the

method given by the present specification (two methods using the composite

efficiencies Of 0.5 and 1.0), the group pile method, and above-mentioned linear interpolation method giving a total of four methods. The comparison of these are shown in

Figs. 7-1 and 7-2. In Fig. 7-1 the horizontal force is 160 tons and in Fig. 7-2 it

is 25- tons. Apart from these values, the analytical values are compared with

the measured values for the four cases. From such results, it can be said that the

linear interpolation method produces values considerably close to the measured values.

7-2 Analysis, With Consideration of the Relative Displacement of Piles

(i) Derivation of the Basic Formula

A model is assumed as shown in Fig. 7-3 to represent the deformation under

a load. When the structure is deformed, the plane A-A becomes stepped as illustrated due to the relative displacement of the piles. Here the following

assumptions are made;

(1) The intersections of the plane A' -A' and the center lines of the piles

are on a plane B-B. This makes it possible to treat the entire behavior of

the model in the same manner as a beam subjected to a shear deformation.

(2) The shear deformation of the piles is ignored.

VII-25
The basic formula is derived on such assumptions. If the displacement of

pile at an arbitrary point is represented by u and w, or;

(7-9)

where

:

u , w : Components of displacement in x and Z directions

o o

u ' : Differentiation for Z o

The coordinate system is shown in Fig. 7-4.

The direct strain £ and the shear strain Y are

z xz

(7-10)

The formula of equilibrium is derived from the principle of virtual work as follows

.

^ qSudz — /

~^ P dudz

C " / ( o^dw +-^^ Sw) dA) ' = 0 ^ A

where

q

pX

T XZ

(7-11)

Ground reaction

Distributed external force working in direction

Surface force working in Z and x direction at the sections of both ends

X -direction component of the outward unit normal line vector of the

sections of both ends

Transforming the equation (7-11) , the differential equations of equilibriiam

expressing displacement can be obtained as follows:

- ( EAw[ ) =:- 0

_ {EI^e'Y - OA =0

VII-26
where :

I*: Moment of inertia at the gravity center due to the axial force of pile

Ip: Total of moments of inertia of all piles

If the displacement 9 is eliminated from the equation (7-12) , the following

differential equation of equilibrium with respect to the displacement is obtained;

(7-13)

(ii) Selection of Modulus of Elasticity

The modulus of elasticity E is assumed. 2.1 x 10^ (t/cm2) for steel.

The shear modulus of rigidity is assumed using a model as shown in Fig. 7-5. The

shear modulus of rigidity is obtained using the average elastic shear deformation

of the joints for the entire piles. If the shear deformations of the joint and

the main body of pile are represented by Y ^^d y , respectively, then;

c s

G A

.D

r . =

G A

(7-14)

where

:

G : Shear modulus of rigidity of material of the joint

c G^: Modulus of rigidity of whole pile

The equivalent shear modulus of rigidity G can be determined by the use of

the equation (7-14) as follows;

GA^

(.D + dV

0 Ac c +

G A

(7-15)

(iii) Application of a Simple Case

An application is made relative to an experimental specimen as illustrated

in Fig. 7-6. The joint is grouted with cement paste, because the model is

small is size. There are two models, one of them uses three piles and the other

six piles. The model is assigned to be placed in the air, and the pile heads are

hinged and the bottoms are rigidly connected and a horizontal load is applied to

the head.

In obtaining the general solution to the equation (7-13) , if the spring

constant of the ground k is o and the x-direction distributed external force Px

VII-27
is 0, the equation (7-13) can be rewritten as follows;

u. (6) 0

(7-16)

P

The general solution to the equation (7-14) is

(7-17)

The boundary condition for the hinged head is obtained by assuming it to be

in the vertical and horizontal directions and rotation. The angle of rotation

at the end of pile and for the pile as a whole are not the same, because the

piles are displaced in relationship to each other. From such boundary conditions,

the unknown constants C can be obtained by solving the six simultaneous

1 5

equations given in Table 7-1.

(iv) Calculation Results

The displacement of the pile head and the stress in the pile body were

calculated by varying the shear modulus of rigidity of the joint. The following

moduli of rigidity were used.

(G) Gc = Gs: The equivalent shear modulus of rigidity, when the section of the

(G) G = 0° : Infinite shear modulus of rigidity, that is, the shear modulus of

(equivalent G) : The shear modulus of rigidity obtained by method (iii)

In Fig. 7-7 the ratio of (equivalent G) to (G) Gc=Gs is plotted on the

horizontal axis and ratio of (y ) to (y ) G = «> on the vertical axis. If the

o o damping rate for Gc (shear modulus of rigidity of the cement paste) is taken

at 1.0, the abscissa would be 0.45. Also y /{y ) G = °° would be 1.02, 1.075,

o o 1.080 and 1. 375 when N (number of piles) = 3 (5,=1.615m), N-3(S,=0.7m) and N-6

()l=0.7m). The bond rigidity of the cement paste and steel material is taken

as 0.3 of the shear rigidity of the cement paste. Then (equivalent G) / (equivalent

G) Gs=Gc would be 0.19. Thus, from Fig. 7-7, y /(y ) G = °° would be 1.025,

o o 1.110, 1.125 and 1.575, respectively.

7-3 Analysis with Consideration of the Bending and Shear Deformations

(i) Derivation of Basic Formula

A model is constructed as shown in Fig. 7-8, which represents a sheet pile

foundation structure. Assuming the entire piles to be an integral beam, the

joint is assumed to be made entirely with steel

rigidity when there is no displacement of the piles in relationship

to each other

VII-28
basic formula is derived by taking into consideration the bending and shear deforations

.

The displacement of pile at an arbitrary point

W — W — d X

(7-18)

The direct strain e and the shear strain Y can be obtained by use of the

z 'xz equation (7-16) as follows

r^^-=u/-e ] (7-19)

From the principle of virtual work, the equation of quilibrium can be obtained

as follows

L / (0^ 5^-+ r^^5u) dA] rno

(7-20)

From the equation above, the differential equations of equilibrixim expressing

displacement can be obtained as follows;

iEAw^') 0

(-Eie') ' -ga e ) (7-21)

If the displacement 9 is eliminated from the equation (7-19) , the following

differential equation of equilibrium in relation to the displacement is

obtained;

El

'J A (7-22)

VII-29
(ii) Application of a Simple Case

Application of these equations is made relative to an experimental specimen,

as illustrated in Fig. 7-6. The shear rigidity of the joint is determined by

use of equation (7-15) . If, in equation (7-22) , the spring constant of the

ground k is 0 and the x-direction distributed external force is 0.

y = 0 o

(7-23)

The general solution will be

y = C, + C-Z + C.Z^ + C.Z^

O 1 2 J 4 (7-24)

However, the unknown constants ~ can be obtained from the boundary conditions as follows;

C -

I 3£/

+

^3=0

2EI

_ P

6E1

GA

OA

(7-25)

(iii) Calculated Results;

In Figs. 7-9 and 7-10, rj ' Gc ( t; is damping rate) is plotted on the

horizontal axis and the horizontal displacement of pile head on the vertical

axis . A comparison is then made of these results by taking into consideration

the displacement of piles in relationship to each other. Fig. 7-9 shows a case

in which there are six piles and the pile length is 1.615 m. Fig. 7-10 shows

a case in which there are the same number of piles (6) , but the pile length is

0.7 m. From these results, it may be said that the two methods of analysis

produce little differences, relative to the horizontal displacement of pile

heads

.

Fig. 7-11 shows a comparison of the stress distribution, as obtained by the

two methods of analysis. It can be said that the two methods procude great

differences with regard to the distribution of stresses. Examination of the

distribution of stresses measured in the in-site tests, it may be said that the

analysis when taking into consideration the displacement of piles in relation

to each other shows a better approximation.

VII-30
8. The Problem to be Solved in the Future

The method of design of sheet pile foundations has been established. However, this

type of foundation can have a wide range of dimensions, thus the design of the structures

using the existing method may not be adequate to cover such a wide range of designs of the

sheet pile foundations. Studies of the structural characteristics of sheet pile foundations

have just been initiated with an ultimate view of establishing better design standards.

The following subjects of study are being considered;

(1) Structural Characteristics of a Sheet Pile Foundation as a Unit

(i) Construction of mechanical models and studies of the theoretical formulas

(ii) Comparison of the theoretical formulas and the results of loading tests

(displacements, stresses, spring constants, damping)

(iii) Experimental investigation of the ultimate yield strength and rupturing

condition (including torsion)

(2) Characteristics of Some Specific Parts of Sheet Pile Foundations

(i) Evaluation of the rigidity of the joints

(ii) Investigations of the joint connections of the top plates

(3) Cofferdam Type

(i) Investigations of the stresses in the cofferdam

(ii) Investigations of the weldability and strength in the presence of residual

stresses

VII-31
References

1. Sheet Pile Foundation Research Committee: Specification for Design and Construction

of Sheet Pile Foundations, January 1972.

2. Keiichi Komada: The History of Development and Present State of Sheet Pile Foundations,

Foundation Work, March 1975.

3. Susumu Kurata and Fumio Shima: Study on Workability and Structural Characteristics of

Sheet Pile Foundation, Bridges and Foundations, Vol. 4, No. 9, September 1970.

4. Earthquake Resistant Technological Development and Research Committee: Report on Research and Development on Earthquake Resistant Technology, National Land Development

Technological Research Center, March 1974.

5. Earthquake Resistant Technological Research Committee: Report on Research and

Development on Earthquake Resistant Technology, National Land Development Technical

Research Center, March 1975.

6. Keiichi Komada, Asao Yamakawa and Michio Okahara: Structural Characteristics of Sheet

Pile Foundations (Report NO.l ), Civil Engineering Technical Material, VOL 17, No. 3 February 1975, Civil Engineering Research Association.

7. Compiled by Iwao Yoshida: Designing of Piled Foundations, Kensetsu Toshu, April 1974.

VII-32
T;»ble 2-1 Dimensions & weight of steel pipe pile

Outer Thick- Section Unit Moment Section Radius Outer

M O O C? area. weight OI modulus of gyra- suriace

meter

2

inertia

3

tion of area

mm mm cm kg/

m

I cm2 Z cm area i cm m^ / m

9 167. 1 131 730xl02 243x10 20, 9 1.88

12 221. 7 174 958x10^ 320x10 20. 8 1. 88

600

14 257. 7 202 111x10^ 369x10 20, 7 1, 88

16 293. 6 230 125x10-^ 418x10 20, 7 1. 88

9 195.4 153 117x10^ 333x10 21. 1 2, 20

700

12 259.4 204 154x10^ 439x10 21. 3 2. 20

14 301. 7 237 178x10^ 507x10 21, 3 2. 20

16 343. 8 270 201x10-^ 575x10 21, 2 2. 20

9 227. 3 178 184x10^ 452x10 23. 4 2. 55

12 301. 9 237 242x10^ 596x10 28. 3 2. 55

O 1 o o 14 351. 3 276 280x10-^ 690x10 28. 2 2. 55

16 400. 5 314 318x10-^ 782x10 28, 2 2. 55

9 340. 2 267 346x10^ 758x10 31, 9 2. 87

12 396. 0 311 401x10^ 878x10 31. 8 2. 87

914. 4 1 A 14 4d1. o

Q C /I /I yi c? 1 r\3 44dx1U 997x10 0 1. 8 2.8 7

16 534, 5 420 1170x10^^ 117x10^ 31. 7 2. 87

9 378. 5 297 477x10^ 939x10 35. 5 2. 19

12 440, 7 346 553x10^ 109x10^ 35, 4 2. 19

1016

14 502. 7 395 628x10"^ 124x10^ 35, 4 2. 19

16 595. 1 467 740x10"^ 146x10^ 35. 2 2. 19

VII-33
(M

0)

1-4&o

x:m

a>

-(->

(1)

a>

(L> +->

-a

(Ua

OS X!

COI

II!

CO

Q)a

nJ

CO I—I aua

(Mt CM

42

Section

modulus

cm"^

N

o

in

00

2,

670

17,

000

X 1—1 54.

800

Weight kg/cm 1

One piece 18. 0

Section

area cm One

piece 22.

99

Dimensions

mm

CO

+->

CM

s: CO

CO

CO ;uiOf

Weight kg/m One

sheet

o

CO

t—1

Section

area cm^One

sheet 165. 0

Dimensions mm

CM

+-> 13. 5o

t—1

1—1

o

r-t

x>

COo

m

o

CVJ

CO

00

YSPB -74

s

—

IT 1

IL 1 1

CO

Section

modulus

cm

1,

420

Moment

of

inertia

cm'*/ m

180,

000

Weight kg/m

00

aa

S

ScM

•2

S ^ o

a;

502.

CO

Section

modulus

cn

6,

600

Moment

of

inertia

cm^/m

152,

000

Weight kg/m inCO

area

n CO

Section cm^/i

447.

VII-34
Table 2-3

allowable stress intensity ( kg/cm^ ) SS 4 1 S S 5 0 SM 5 0 A

1 . axial tensile stress intensity

(per all section area)

1, 400 1, 700 1,900

2. axial compression stress intensity

(per all section area)

compressive members

A = length of members

T = sectional radius of all cross section

compressive added member

0<^/r^l 10

1.300-0.06(i!/'r)^

(/V>110

7,200,00(^(^/r r

0<^/r^l00

1,600-00 9 (//r)^

^/r>100 , 7.200,000/( l/r)

1. 600

0<^/r^90

1.800-0.1 l(,;/r)^

(/r> 90 , 7,200,000/(/;/r)

1, 800

3. bending stress intensity tensile fibre of beam (per all section area)

compressive fibre of beam

(per all section area)

i = distance between fixed points

of flange (cm)

b = width of flange

in this case beam with being fixed compressive flange directly by

reinforced concrete deck and so

forth

1. 400

1,300-0.6 ( //A )^

wficre c/ 0 ^ ou

1, 300

1, 600

l,600-0.9( l/bf

wh(*rp / /h <C ^0

1, 600

1.900

1,800-1.1 (^/4)^

where //A <1 30

1. 800

4. members for axial compressive load and bending moment

whose members must be calculated by formulas to be given below.

for buckling about horizontal axis

for buckling about vertical axis

P M a ca\ + yc A 1 1,300

< a cat P M — + — yc cai A 1

titi L P

A

3

P M Oral + yc

A 1 1.600

9 <^ 0 r g 1

to left

P .W Oral + yc

A I 1,800

y ^ 0 ra 1

to left

Table 2-4

basic type section shape number of

foundations

treatment of

joint

well type

pronged

type

circular oval

oval rectangular

other

single

complex

grouting mortar

welding

no treatment

VII-35
Table 3-1

kinds of piles

kinds of

times in loading

bearing

pile

friction pile

good sandy

stratum

other

normal 3 3 4

erarthquake 2 2 3

Table 3-2

1) well 2) pronged

e

Av

R =

7l| • S

i) distance between each

prong over 2. 5D

ii) distance between values smaller

each prong between (1) and

under 2. 5D (2)

(1) same to 2) i)

1 f-U'l,

(2)

1 q^'A^ + f-U'd.+l^)]

VII-36
Table 3-3 Composite efficiencies

Kinds of

sheet piles Treatment of joint Composite

efficiency

Remarks

Hshaped

Filling with concrete in part

of flange with dowel

connectors

1. 00

Filling with concrete in all

section

1. 00

Filling with concrete in part

of flange 1. 00

Welding 50 cm part of joint 0. 40

Fixing the head 0. 40

wm

Steel

pipe

Grouting mortar and fixing

the head 0. 50

Fixing the head 0. 30

Table 3-4

modulus of deformation of ground by

the following tests Eq (kg/c.m^)

a Normal earthquake

a half of the Eq obtained by the plate

loading test using a 30 cm diameter

disk

1 2

the Eq measured inside a boring hole 4 8

the Eq obtained by the unconfined

compression test or the triaxial

compression test

4 8

the Eq estimated by Eq=28N;- the

N-valve obtained by the standard

penetration test

1 2

VII-37
ooiH

»-> Unu

+jcouo

*-» aac

.2

-f-> a>

(U X!

cao

(0

(U -aa

aJXHI

I—I a

0)a

0)

6)

>-> m

0)00

•iH

J3

O

ti

o

c—' o o

a; .S a

c

S

c a, .2 °- w 'a,

fl) f—' c <u

•tH •4-' T3 M

!2

01

T3 to o Z

Sia

uO

0)

T3

0)Cou

a. oo

o'

Q.C

03O

CO

O

O

r;

o

o ^ ° o • w CO

<M

CO

CO

o"

CMX

r-. to o

T3

•r-(

CIJ OXo

o

nj rt

0) ^ U M o o

n!

03

OO

CM ^ O

O

o

o

CM

O CM

in

C7> O

in

<*<

ni Ucu

cd

rtbo o

c

0)

•rH oo

C^^

irT COX

CM

t-HX

CO

cm'

CO

«1< t-co

COX

CO

in

0)Co

Q. oo

r-o'

in

CO

oo

r;

oo ^ o ° CO ^ _

/v," <^ ^ ^ in CO

in

CM*

in

n)

03 o

03

03 oD

03coc

03

ooo

co" CMC

03oooo

<r

CDX

1—

t X

03Xo

en'

inoc

nJ Z

ooC

03 Ooo

o'

CMXX

CM

COo

t-' X

CO

CMCO

COu

03

n)O

rt

rtC

I/)

oooo

CM

CM

CO

cmc

"(-, ooooo

CO

COX

COX

05

oo

in

CO

COX

CO

t-lX

CD

o»-) CM

in

o

CM

in

COXo

in

CO

CO CO

CD

03 o

com

03 ? ,3

a

cs -ir 3 o a a; fc u o

(X

03 O

VII-38
basic type

well

pronged

well

pronged

well

length of

joint

14.

000

12.

500

33,

700

22,

700

26,

000

16,

000

6,

500

33,

000

30,

000

type type

of

joint pipe

pipe-T pipe

pipe-T

type

coffering

no

closing

coffering

rising

m

number of foundation to 03

1—) CO (N rH 1—1 CO iH

Dale

of

work

dimensions

of

steel

pipe

pile

762x9x27,

000

914.

4x12,

7

x24,

000

914,

4x12x35,

000

914.

4x12x28,

870 O O

o ir>

1-t lO

(N

1-) o

o

CO

914.

4x12.

7

xl3,

900

914.

4x14x34.

500

800x12x32,

000

m

basic

sectional

dimensions

6

13,

000

6

12,

936

6

12,

726 13,946x5,811

6

9,

081

12,

560x9,

620

10,

600x7,

660

12,

110x4,

010

7,

581x14,

527

15,

727x20,

376

11,

658x3,

710

date of

start 46.

12

(1971) 46.

12

46.

12

46.

12

47. 7

47. 947.

10

»-)

!>'

•>«•

47.

12

48. 1

location

of

work

(prefecture)

Okayama

Tokyo Saga

Miyagi

Okayama

Osaka Miyagi

orderer

Okayama

Highway Public

Corporation

Tokyoto

Bureau

of

Kyushu

Bureau

of

Tohoku

Miyagiken

Okayamaken

Osakashi

Miyagiken

name

of

work

(bridge)

Yoshii

Mizushima

Saidaiji Ariake Quay

Hazenoura

Shiniinogawa

Shinkitakami

Gokeidai Shibatani

Hirano

Line

Shinkitakami

VII-39
basic type

- —

pronged ~

well -

gth joint o

o

t- o

o

r- o

o

r- o

o

in o

o

r- o

o

1/5 o

t~ o

o

CM o

o

CM o

o

r~ o

o

r~

O

CO

CM co

CM CO

f-H o in

CM 00

T-H CM o

CM

toin

CO

CM COCM

Eh

1 Eh

type type

ioin pipe pipe pipe pipe

type

coffering -

in

water

coffering z =

—

no

closing

in

number of

foundation CM CM CM CO rH in in 00

scale

of

work

dimensions

of

steel

pipe

pile

1,

200x14x28.

000

914.

4x14x39,600

800x14x20,

000

508x12.

7x11,

000

800x16x27,

000

800x12x20.

000

1,219.

2x16

x31,

000

914,

4x16x21,000

1.

219.

2x19

x57,

500

914.

4x12x28,

000 o

o

o

o"

COX

<i< e<i tH iH X

CO

rHo

fH

basic

sectional

dimensions

22,

225x15,

008

6

24,

508

6

15,

488

6

3,

910

6

15,

506

6

10,832

10,

568x25.

193

6

9.

081

26,445x16,

186

CM

t~00

to' -€>

14,

062x10.

504

other

date of start 48. 2

(1973) 48. 3

48. 5

48. 5

48. 6

48. 748. 948.

11

48.

11

48.

11

48.

12

location

of

work (prefecture)

Osaka

Tokushima

Hokkaido

Hyogo

Hokkaido

Hyogo Aichi Chiba

Nagasaki

Chiba

orderer

Osakashi

Tokushimaken

Bureau

of

Hokkaido

Nippon

Steel

Corporation

Bureau

of

Hokkaido

Development

Kobeshi

Aichi

Bureau

of

Kanto

Bureau

of

Kyushu

Bureau

of

Kanto

name

of

work

(bridge)

Minami

Quay

Connection

Suehiro Ishikari

Kurogane

Shinebetsu

Teshiokako

Rokko

Island

Connection

Kinuura Shinsuigo

Kofunadu

Ichikawa Waterway

3

VII-40
basic type

well

type

length of

joint

41,

000

12,

700

34,

700

26,

500

41,

000

6,

000

30,

000 o

o

00

lo"

type

of

joint pipe

pipe-T pipe

second port type

1

pipe

(1) Q.

+^

cn

coffering

no

closing :

coffering

no

closing

coffering

-

scale

of

work

number of

foundation CO i-< CO

dimensions

of

steel

pipe

pile

914.

4x14x44,

000 o oo o

o o

cd" •*" CO r-l X

-r N

»-H I-) X! CO

»-t o

TH

o

o

o

00

COX

CO X

CO

1-t o

1-1

1,219.2x16

x42,

500

508x9x6,

000 rCD CO

i-( rH O

X o

CO o

cJ o"

iH CO

CO X

914.

4x16x38,

000

1-1

basic

sectional

dimensions

6

18,

684 15,200x11,700

13,

100x9,

700

26,

058x14,

322

2,

000x3,

954

6

8,

no 6

8,344

32,

897x12,

422

35,

288.

5x13.

35

date of 49, 3

(1974) 49. 3

49. 3

49. 5

1-1 »H

OJ

CO

1-t

50. 1

location

of

work (prefecture)

Tokushima

Chiba Chiba Osaka Hyogo

Kanagawa

Ishikawa

orderer

Bureau

of

Shikoku Bureau

of

Kanto

Osakafu

Bureau

of

Kinki

N.

K.

Steel

Corporation

1

Ishikawaken

Hanshin Highway Public

Corporation

name

of

work

(bridge)

Shinkagasuno Edogawa

Bridge

Left

Site

Edogawa

Bridge

Right

Site

Senboku Yaegawa

Side

Road

Ogishima

Jonan

Osakashi

Highway Osaka

Gulf

Circum- ference

VII-41
13

o

.s

4)a

m

co o 4S c o c

3 3

Cm

C nj O Q,

c I*

•a 03

o

cn

05c

41

.'n a

0)

n! o

T3 01

in

CO

o u

c ^ 2

O (ua

o

o

01 •ouo

-r-t O u

rtC O

IM

COo

1-H (M

in"

i-tX

tDO

i-H pa

I in

1—1

urement

stress o o o o o

g

c o o o o o o

>

bJO

loadir test o o o o o

of

bridge

Kaneshiro

nogawa

Ltakamiga

Mizushima )

Connect]

loura rt& 0

Ishikarikako

o

W)

•iH

name

Nankc Shinii

Shink; Nankc Hazer Omig;

Suehi;

Shinsi

I

a;

1-1

-s

Eh

VII-42
basic type well well well well

length of

joint

17,

400

15,

300

type type

of

joint pipe pipe pipe pipe

type Foundation

Foundation

Foundation

scale

of

work

number of

foundations CM T-< CM 00

t-(

dimensions

of

steel

pipe

pile

1.

219.

2x12

x22,

000

1,

219.

2x7.

9

x23,

000

other

1

762x7.

9x24,

000

762x

xl6,

000

1,

117x6x21,

000

basic

sectional

dimensions

6

6,

458

6

35,

402

12,

640x29,

072

other 12,

640x29,

072

6

5,

784

12,

700x31,

700

6

13,

824

7,

000x20,

100

6

15,

600

date of

start 41. 1

(1966) 41. 6

44. 5

1

44. 9

45. 1

45. 145. 345. 9

location

of

work (prefecture)

Okayama Okayama Okayama Okayama

orderer

Kawasaki

Steel

Corporation

name

of

work

In

Steel

Factory

lst~4th

Blast

furnace

Scale

Pit

•rH (X

Rain-quay

In

Steel

Factory Quay

VII-43
basic type

pronged

well

pronged

-t-> -c .5 & 0

o

o

o

o

ro

o

o

o o

o o

o o

c O CM

co

CM

CM

CM 00

CM CO

type type

of

joint pipe

pipe-T

Second port

shaped

type

Foundation

number of

foandations t-H CM

ale

of

work

dimensions

of

steel

pipe

pile

762x11x29,

000

762x7.

9x23,

000

1,

016x12.

7

x55,

000

812.

8x19x22.

000 O Oo o

o o

CM 00

CM CO

cn

i-<X

CO

CM

T—

t

00

OS

basic

sectional

dimensions

i

5,

784 o<*<

in 15,

248x6,

946

19,

992x15,

248

9.

127x9,

127

6

8,

752

date of

start 46. 5 T-H £—

'a*

48. 9

(1966)

y-l

location

of

work (prefecture)

Okayama

Tokyo

Hiroshima

orderer

Kawasaki

Steel

Corporation

Car

Ferry

Company

Chugoku Electric

Power

N.

K.

Steel

Corporation

name

of

work

Quay

Detached

Pier

Dolphin

Quay Steel

Tower

In

Fukuyama

Steel

Factory

VII-44
basic type

proaged

well well

pronged

length of

joint

32,

500

29,

500

30,

000

25,

000

30,

000

19.

300

type type

of

joint

type

rising rising

number of

foundations CD »—

1

CM •<i< CM

ale

of

work

dimensions

of

H-shaped

sheet

pile

YSPB-74

=48.

500

=39,

500

II

=30,

000

II

=25,

000

tl

=30,

000 II

=42,

800

OS

basic

sectional

dimensions

1,

670x2,

090

5,

770x8,

660

1

4.

930x4,

930

2

4.

530x4.

530

4

2,

530x2,

530

2,

490x2,

066

2,

510x2,

510

4,

200x3.

360

a

4-c of start 41. 9

(1966) 42. 3

42. 8

44. 8

45. 3

45.

10

location

of

work (prefecture)

Aichi Osaka

Kanagawa Okayama

Gifu

orderer

Nagoya

Port

Association

Kansai

Oil

Nippon

Steel

Corporation 1

Keihin

Quay

Public

Corporation Mitsubishi

Oil

Gifuken

name

of

work

Kaneshiro

Dolphin

Takanashigawa

Intersection

Fukusoku

VII-45
(B)

L-75x75x9

T-150X42X12

65

1 r

L-lOOxlOOxlO

125x625x9

circular oval rectangular

Fig. 2-1 Joint shapes of steel

pipe pile

Fig. 2-2

(1) rising type (2) closing type (3) coffering type (4) semi-rigid type (5) rigid type

0 steel pipe pile (D steel pile (if necessary) (3) the inside filled reinforced concrete 0 footing

(?) bottom plate (§) double closing Q) cofferdan'.

Fig. 2-3

VII-46
guide ring

prepacked

concrete timbering

(D driving steel ©

pipe piles

parts of

coffering

foundation body

(4") footing

pier

ofSts"* \ ^^bering

I- prepacked "dry up

concrete

J

cutting

y in water

cutting steel pipe piles in water ® completed

Ho

^1

earth surface

for designing

mil

m

m

Kv

m

m

m

Km

Fig. 2-4 several steps in construction

of the "cofferdam" type of

the sheet pile foundations

Fig. 3-1

30

5 20

10 m

nTable 4-1

I I " 4-2

40

30

c

oi 20 +

•oe

3o

10

5 10 15 20 25 30 35 40 45(m)

ill Table 4-1

" 4-2

0 5 10 15 20 25 30 35n40(m)

Fig. 4-1 Largest width of Fig. 4-2 Diameter of circular

sheet pile foundation sheet pile foundation

reclangulor or oval

VII-47
c

oo

c3

o

tt.

60

50

40

30

20

10

da Table 4-1 O " 4-2 ^ " 4-3

1

s

o

•a

CO •ac

3o

50

40

30

20

10 -

Ii

If

m Table 4-1

" 4-2

1^

0 5 10 15 20 25 30 35 40 45 50 55 60 65 70 75 80 0 200 400 600 800 1000 1200 1400

Fig. 4-3 Sheet pile length Fig. 4-4 Diameter of

(including length of steel- pipe piles

coffering)

VIIn48
12,500

o

o

CM

-8,500

o o

o o

o o.

CO

Tf

o> 3,600

o

o

o Q. o

o

cm'

-12,000

3,500 3,600

-20,000

^ -32,000

-50,000

30,700

9,0 95

o

o

o

to a o

Fig. 5-1 Cross section Omigawa Bri

VII-49
Depth Soil Relative density N-value

0 10 20 30 40 50 60 70

— 450

— 1

Clay

-1246

f—r r-r7-7j-ri-r-. '/^-'^'^

'"7"/ r / /~7~

Silt containing ssnd

-14.71

Fine sand

-1791

Fine sand

-2206

Mediumfine sand

-2721

'-.o

. o: .*.•-'.. . Sandy

gravel Medium sand

containing

gravel

/

f-f—

\

-28.21 -.*.•::*.*.

-32.81

Mediumflne sand J

t-—

-37.91

'^^^^

ji^ y///./

Fine sand

containing

silt

-40.51

'/////' n

-44.75

'/// 'y- •'

OA'//,

n

-46.71 Sandy silt

Silty fine sand -<

Fig. 5-2 Soil conditions; Omigawa Bridge

VIItSO
o

O-^ O O O O '"-H w o ui o w [V, M "

PBO] [C^UOZUOH

peoj [cjuozuoH

VII^Sl
Fig. 5-6 Stress distribution

VII-52
Frequency / (Hz)

Fig. 5-7 Horizontal response on the top end

VII-53
1.0

(b)

_l \ I I I

I 2 4 6 8 10 12

Frequency / (Hz)

Fig, 5-8 Vertical response on the top end

VII-54
contact gauge point

fo

- o

!«

tX

O

»

K

J o

a o o

OUi o O oo oCsJ K O

1 o

«o

1 ^ o

g 16& « t 7.1 « t 1.000

Hi

IbO 247.2 150 24 7.2 [l 50

Fig, 6-1 Direct shear test

b ) :a

Fig. 6-4 Combined beam

fracture

mortar

C-T type pipe type

Fig. 6-2 Fracture of joint

°^00| ^ l.OOU I

1000 : I OOP I ICOO 'MO'?

, 5 OOP

dial gauge

contact gauge

strain gauge

Fig. 6-6 Bending test

t60

100

50

Theoreticil>/=?/,'?Aij'2

1^Experimental connder-,

\iiig footiag rigidity/^ / ^ /

0 5 10 15 20

Central deflection (mm)

ISO

100

50

No. 1 No.lO

1 800 I

, 1800 I j TOn 5000 1

No. 1 No.lO

0 02 04 0.6 08

Denection in the axial direction iT'i mm)

Fig. 6-7 P- 5 curve Fig. 6-8 P- 5' curve

VII-55
S-S section

ISO 120110 100 60 60 bO 40 30 2010

10 2 0 30 AO 60 GO 60 100110 130160 Concennated load P; ton

Fig. 6-9 Load-Strain distribution

1.0 —

« 0.5

From P-e curve From P-i curve

100 p (t

)

150

980 21002400 2730 Measured stress trnfx ^^s/^'""'

6.0 6.0 r'iKjrA"!')

Fig. 6-10 <^ -P curve Fig. 6-11 cc^t' curve

0 tAOOMeasured /i = 0.5 /('l.O

value Present specification

Group pile oi = O.d 5 method Linear interporation

method

Fig. 7-1

VII^56
(^P-12.0)

H = 250

-^00 0 +.400- 0 taOO- 0 +A00- 0 1-aOO- 0 tAOO-'^y cm' —i

' -

/ - —

3^18) J —

1

f

,—

-

- — —

1 /

rMeasured /i = 0.5 U = 1.3 value Present specification

Group pile a = O.Y'f method Linear interporation

method

Fig. 7-2

Before deformation After deformation

Fig. 7-3 Structural model

Centroid of sheet pile foundation

A

Fig. 7-4 Coordinate

system

VII-57
D+ d

Fig. 7-5 Model of shear deformation

Fig. 7-6 Model for calculation

II o

o3V—

-

O

O

o

Ratio of (Equi-G) to (Equi-G) Gc=Gs

Fig. 7-7

VII-58
Before deformation After deformation

Fig. 7-8 - Structural model

Fig. 7-9

VII-59
VII-60
ON SPECIFICATIONS FOR EARTHQUAKE-RESISTANT DESIGN

OF HIGHWAY BRIDGES (JANUARY, 1971)

by

Kenji Kawakami

Director, Public Works Research Institute

Eiichi Kuribayshi

Chief, Earthquake Engineering Research Section

Public Works Research Institute

Toshio Iwasaki

Chief, Civil Engineering Section, USEE

Building Research Institute

and

Yutaka lida

Research Engineer, Earthquake Engineering Research Section

Public Works Research Institute

ABSTRACT

This paper presents the details of new specification for Design of Highway Bridges

to Resist Earthquakes, completed in January 1971 by the Highway Bridge Committee of the

Japan Road Association.

Key Words: Earthquakes; Highway Bridges; Seismic Providlons; Specifications; Structural

Engineering.

VIII-1
Introduction

In the event of a major seismic disaster, highway networks should perform their duties

to facilitate evacuation, rescue and recovery operations with a minimum loss of life and

property. In such a case, bridges, viaducts, etc. which constitute important parts of

highways, would decide the fate of each route of a highway network. Therefore, the structural safety of these structures, against earthquakes, are extremely important.

In 1966 the Ministry of Construction commissioned the Japan Road Association to draw

up a new comprehensive specification for the earthquake-resistant design of highway bridges,

and accordingly provide maximum protection against earthquakes at minimum cost. In response

to the commission, the Japan Road Association established a special committee to develop

these new specifications, which were completed in January 1971. The Japan Road Association

immediately reported to the Ministry of Construction, who then notified the highway administrative organization of these specifications. These specifications were then issued as

directives of the Road and City Bureau on standards for bridges and viaducts in March 1971.

The directives issued are supplements to Article 5 of the Ministry of Construction

Ordinance for the enforcement of the Road Structure Ordinance, which is a governmental

ordinance. The legal conditions of this ordinance is as follows; The Road Structure Ordinance is in force and conforms with the Road Law. Article 30 of the Road Law provides that

technical standards for bridges and other principal structures, which are designated by a governmental ordinance, may be subjected to a governmental ordinance. Paragraph 1 of

Article 35 of the Road Structure Ordinance provides that bridges, viaducts, etc., shall

be constructed of steel or concrete, of similar nature. Paragraph 2 of Article 35 provides

the total design live loads shall be 20 or 14 tons, for bridges, etc. Paragraph 3 states

that other matters concerned with the structural standards of bridges, etc. shall be provided by a Ministry of Construction ordinance. Article 5 of the Ministry of Construction

Ordinance for the enforcement of the Road Structure Ordinance provides that bridges , viaducts

,

etc. shall be sufficiently safe against dead loads, live loads, wind loads, seismic effects

and other loads which are expected to act on bridges and any combination of loads, considering structural types, traffic situations, topographical, geological, meteorological and

other conditions.

Outline of Specifications for Earthquake-Resistant Design of Highway Bridges (January 1971)

Since 1956, the earthquake-resistant design of highway bridges had been conducted in

accordance with the provisions of Section 13 (Section 2.9 in the English version) of

"Specifications for Design of Steel Highway Bridges" which were issued in 1958 by the

Japan Road Association. These specifications were revised in 1964 (the English version

was issued in 1968) and subsequently displaced by "Specifications for Highway Bridges"

in 1972, the details of which are as follows;

(1) The horizontal design seismic coefficient shall be determined by Table 1.

(2) The vertical design seismic coefficient shall be assumed equal to 0.10.

(3) Increases in the allowable stresses for only seismic forces or for dead loads plus

seismic loads shall be taken as

VIII-t2
70% for steel structures

50% for concrete structures and reinforced concrete sturctures

Since the provisions in the Section 13 were not comprehensive, it was necessary to

draw up new detailed specifications exclusively for aseismic design of highway bridges.

Current "Specifications for Earthquake-Resistant Design of Highway Bridges" were issued

in January 1971, by the Japan Road Association, which apply to the design of highway

bridges with spans not longer than 200 meters, to be constructed on expressways, national

highways, prefectural highways and principal municipal highways.

The specifications basically require that seismic coefficient methods be applied

and that methods be used in determining the seismic coefficients. One method is the conventional seismic coefficient technique that applies to the design of relatively rigid structures. The other method is the modified seismic coefficient technique considering the

structural response that applies to the design of relatively flexible structures. The

following lists the principal points of the specifications;

(1) The horizontal design seismic coefficient for a rigid structure is determined systematically, depending on the geographical location of the bridge site, the ground conditions at each sub-structure site, and the importance of the bridge. The horizontal design coefficient, for a flexible structure is determined as a function of the fundamental natural period of each structural system.

(a) In the seismic coefficient method that is employed for relatively rigid structures,

the horizontal design seismic coefficient (k, ) shall be determined by

h

k = v V V k (1) h 12 3 0

where :

k : horizontal design seismic coefficient,

h

k : standard horizontal design seismic coefficient (=0.2), o

v^: seismic zone factor,

v^ : ground condition factor

,

v^: importance factor.

The values of v^, v^ and v^ are shown in Tables 2, 3 and 4 respectively. The

definitions of classification are specified in the provisions. The minimum value

of k, shall be considered as 0.10.

h

(b) In the modified seismic coefficient method considering structural response that

is employed for relatively flexible structures such as a bridge with highrise piers

higher than 25 m or a bridge with a fundamental period longer than 0-5 seconds,

the horizontal design seismic coefficient (kj^jjj) shall be determined by

\m ^kh

VIII-3
where :

k : horizontal design seismic coefficient in the modified seismic coefficient hm

method considering structural response

k. : coefficient given by eq. (1) h

3 : a factor dependent on the fundamental period of the bridge , and obtained by

Fig. 2. For structures whose fundamental periods are shorter than 0.5 seconds

3 may be considered as 1.0.

The minimum value of k, shall be 0.05. hm

(2) The vertical design seismic coefficient may generally be considered as zero, except

for special portions such as bearing supports.

(3) The horizontal design seismic coefficient for structural parts, soils and water below

the ground surface may be considered as zero.

(4) Hydrodynamic pressure during earthquakes are specified in the specifications.

Earth pressures during earthquakes, however, are specified in the related specifications .

(5) Specific attention should be paid to very soft soil layers and soil layers vulnerable

to liquefaction during earthquakes. The bearing capacities of these layers are to be

neglected in the design, in order to assure high earthquake-resistance for structures

that are built in these layers.

(6) Special attention should also be paid to the design of structural details, in

consideration of the damage previously experienced to bridge structures. To this aim

provisions are specified for bearing supports and devices for preventing bridge

girders from falling.

(7) Increases in the allowable stresses of materials may be considered in the earthquakeresistant design, magnitudes of increases for various materials are specified in

several related specifications, and the increasing rates are as follows;

concrete in reinforced concrete structures; 50%

reinforcements in reinforced concrete structures: 50%

structural steel for superstructures: 70%

structural steel for substructures! 50%

concrete in prestressed concrete structures against

compressive forces:

foundation soils: 50%

VIII-4
Acknowledgrtients

Specifications for Earthquake-Resistant Design of Highway Bridges (January 1971) are

the fruit of efforts of the Earthquake-Resistant Design Branch-Committee of the Specifications Subcommittee under the Highway Bridge Committee, the Japan Road Association.

Special thanks are extended to the Branch-Committee headed by Mr. Yasuo Tada for his

energetic efforts.

VIII-5
Table 1, Horizontal Design Seismic Coefficient (Out of Date)

Ground

Conditions*

Regions*

Weak Ordinary Firm

Where severe earthquakes

have been frequently

experienced

0. 35-0. 30 0. 30-0. 20 0. 20-0. 15

Where severe earthquakes

have been occurred 0. 30-0. 20 0. 20-0. 15 0. 15-0. 10

Other regions 0. 20 0. 15 0. 10

*no further description on regions and ground conditions.

Table 2. Seismic Zone Factor for General Highway Bridges

Zone Value of

A 1. 00

B 0. 85

C 0. 70

Note: Refer to Fig. 1.

VIII-6
Table 3, Ground Condition Factorvg for General Highway Bridges

Group Definitions ^ ^ Value of

1

(1) Ground of the Tertiary era or older

(defined as bedrock hereafter)

(2) Diluvial layer2) with depth less than

10 meters above bedrock

0. 9

2

(1) Diluvial layer^) with depth greater

than 10 meters above bedrock

(2) Alluvial layer^^ with depth less than

10 meters above bedrock

1. 0

3

Alluvial laver^^ with deoth less than

25 meters, which has soft layer^) with

depth less than 5 meters

1. 1

4 Other than the above 1. 2

(Notes) 1) Since these definitions are not very comprehensive, the

classification of ground conditions shall be made with

adequate consideration of the bridge site.

Depth of layer indicated here shall be measured from the

actual ground surface.

2) Diluvial layer implies a dense alluvial layer such as a

dense sandy layer, gravel layer, or cobble layer.

3) Alluvial layer implies a new sedimentary layer made by

a landslide.

4) Soft layer is defined in Section 3. 7 "Soil Layer Whose

Bearing Capacities are Neglected in Earthquake Resistant

Design".

Table 4. Importance Factor for General Highway Bridges

Group Definitions Value of

1

Bridges on expressways (limited-access

highways), general national highways and

principal prefectural highways.

Important Bridges on general prefectural

highways and municipal highways.

1. 0

2 Other than the above 0. 8

Note: The value of niay be increased up to 1. 25 for special cases

VIII-7
Nolural Penods

ZD

T lied

Fig. 2. Factor

VIII-8
APPENDIX

SPECIFICATIONS FOR EARTHQUAKE-RESISTANT DESIGN

OF HIGHWAY BRIDGES (JANUARY 1971)

Tentative English Version

JAPAN ROAD ASSOCIATION

VIII-9
FOREWORD

The earthquake-resistant design of highway bridges is currently being conducted in accordance

with the provisions of Section 13 (Section 2.9 in the EngUsh version) of the Specifications for Design

of Steel Highway Bridges. Since the provisions concerning earthquake-resistant design are not very comprehensive as stated in the preface of the Specifications, the design is in many respects dependent

on the designer's interpretation. Accordingly the results of the various designs were not always

consistent. To alleviate this problem it was necessary to revise the Section 13 of the Specifications

and to draw up new detailed specifications exclusively for earthquake resistant design of highway

bridges.

In response to the need for revisions, the Japan Road Association established the Earthquake

Resistant Design Subcommittee under the Highway Bridge Committee in 1966. Between 1966 and

196S a preliminary draft was prepared. The task was taken over by the Earthquake-Resistant Design

Branch-Committee of the Specifications Subcommittee in 1968, when the Highway Bridge Committee

was reorganized.

The Subcommittee and the Branch-Committee have seriously attempted to draw up the revision throagh surveys of the available research work, discussions with specialists from universities and other

org2j:iizations, and trial calculations and bridge designs at several stages in accordance with the interim

drafts, and finally completed the revision of the Specifications in 1971. The new Specifications are completed basically in accordance with the provisions of Section 13 of tiie Specifications for Design of Steel Highway Bridges, and also by extensively referring to the

resu!3 of recent investigations on earthquake engineering. Therefore the design methodology of

SectiDc 13 of the current Specifications was generally traced in establishing the new Specifications. la determining the design seismic coefficient the importance factor is newly considered in

addition to the traditional factors such as zone factor and ground condition factor. Moreover, for brides which have longer natural periods the factor on the natural periods of the structures are

consiiered. Although the design horizontal seismic coefficient used to be 0.1 to 0.35 in the Section 13 of

the icaxrent Specifications, the value is reduced to 0.1 to 0.3 and over 0.3 only for rare occasions in the lew Specifications. This is based on the results of recent investigations that it seems more

esse:;~al to pay attention to soil conditions rather than to the magnitude of seismic coefficients in desjcL of bridges against seismic forces. Saice design seismic forces in the new Specifications were determined by referring to the seismic

risk nap expected for 75 years presented by Dr. Kawasumi and by reflecting the results of a recept stu^v on the meteolorogical data since the Kanto Earthquake in 1923, seismic forces considered in the lew Specifications would cover the severest earthquakes experienced in individual regions in

Jap£2.

Hie seismic forces would also correspond to those due to the greatest earthquakes in the

seismrlogical point of view.

Li the new Specifications the following aspects differ from the previous method.

(1) Th.e design seismic coefficients of a bridge are determined systematically dependent on

Kographical location of the bridge site, the ground conditions at each substructure site, and the jnportance of the bridge.

(2) Toe vertical design seismic coefficients are generally neglected.

(3) Tie design seismic coefficients for structural parts, soils and water below the ground surface are

iBglected.

(4) Saecial attention is paid to very soft soil layers and soil layers vulnerable to liquefaction during

arthquakes. The bearing capacities of these layers are neglected in the design in order to assure ugh earthquake-resistance for those structures which are constructed in these layers.

(5) Tie design seismic coefficients for bridges with highrize piers are determined in accordance with ne modified seismic coefficient method considering structural response, instead of the convenrimal way in which the design seismic coefficients increase with the height of the piers.

VIII-10
(6) Special attention is also paid to the design of structural details in view of earthquake damage

previously experienced to bridge structures. To this end provisions are specified for bearing

supports and devices for preventing superstructures from falling. Although the Specifications were developed through extensive studies and discussions of research reports on earthquake engineering from throughout the world as well as Japan, some unexpected

problems may arise when they are applied to actual structures. Because of this and the fact that technological improvements take place very rapidly, another revision will be required in the future. Everyone is urged to indicate any insufficient points or to make any suggestions for the present

Specifications, so that they may be improved in subsequent revisions. The present publication is hmited to the provisions which are considered to be sufficient for the design of bridges. Comments which dedscribe the basis and the application methods of the provisions

will be issued immediately after their completion. They will be indispensable when it is necessary to

revise the present Specifications again or when any parts of the provisions are found to be uncertain

in their apphcation.

Dr. Hiroshi Kawasumi, Dr. Tsutomu Terashima, and Mr. Tsutomu Tomita participated in the

discussions on special matters as special members of the Committee. Their cooperation is greatly

appreciated.

January, 1971

Earthquake Resistant Design Branch-Committee

Specifications Subcommittee

Highway Bridge Committee

Japan Road Association

3-3-3 Kasumigaseki

Chiyoda-ku, Tokyo

LIST OF MEMBERS

HIGHWAY BRIDGE COMMITTEE

Chairman: Aoki, Kusuo

Members: Arie, Yoshiharu

Hirai, Atsushi

Kokubu, Masatane

Matsuzaki, Akimaro

Nakajima, Takeshi

Naruse, Katsutake

Okumura, Toshie

Tahara, Yasuji

Uemae, Yukitaka

Fukuda, Takeo

Hiura, Taizo

Konishi, Ichiro Murakami, Eiichi Nakano, Takayuki

Okamoto, Shunzo

Ozaki, Hisashi Tanaka, Goro

Yokomichi, Hideo

Tomonaga, Kazuo

Fukuoka, Masami

Inomata, Shunji

Kono, Michiyuki

Murakami, Tadashi

Naniwa, Hayato

Okubo, Tadayoshi

Suzuki, Toshio

Secretaries: Komada, Keiichi Okada, Tetsuo

Kutsukake, Tetsuo

Sawai, Hiroyuki

Miyazaki, Shoji

VIII-11
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XIES
CHAPTER 1 GENERAL

1.1 Scope

The provisions in the Specifications apply to earthquake resistant design of highway bridges with

spans not longer than 200 meters, to be built on expressways, national highways, prefectural highways

and principal municipal highways.

In regard to matters which are not specified herein, the following Specifications shall be conformed to, in accordance with the type of the structure considered.

Specifications for Design of Steel Highway Bridges, Japan Road Association

Specifications for Design of Welded Highway Bridges, Japan Road Association

Specifications for Design and Construction of Composite Beams for

Steel Highway Bridges, Japan Road Association

Specifications for Design of Reinforced Concrete Highway Bridges, Japan Road Association

Specifications for Design and Construction of Friction Type Joints

in Steel Highway Bridges Using High Strength Bolts, Japan Road Association

Specifications for Design of Substructures of Highway Bridges, Japan Road Association

Specifications for Prestressed Concrete Highway Bridges, Japan Road Association

1.2 Definitions of Terms

The following definitions apply only to the provisions of these Specifications.

(1) Earthquake: A phenomenon with propagation of vibration due to a sudden naturally occurring

movement at a certain portion inside the earth.

(2) Earthquake Ground Motion: Vibration of ground during earthquakes.

(3) Design Seismic Coefficient: A coefficient indicating the magnitude of acceleration to be

considered for earthquake-resistant disign of structures, and expressed as a fraction of the

acceleration of gravity.

(4) Seismic Coefficient Method: A method for earthquake-resistant design in which seismic forces are assumed to act on structures as static forces.

(5) Modified Seismic Coefficient Method Considering Structural Response: A method for earthquake

resistant design which is developed through modification of the seismic coefficient method by

considering characteristics of earthquake ground motions, dynamic properties of structures, etc.

for those structures with long fundamental periods of vibration.

(6) Modified Design Seismic Coefficient Considering Structural Response: A design seismic coefficient used in the modified seismic coefficient method considering structural response in which the design seismic coefficient is dependent on the fundamental period of the structure.

(7) Standard Horizontal Design Seismic Coefficient: The Horizontal design seismic coefficient which

applies to a bridge which is located in a zone where severe earthquakes have frequently occurred or where severe earthquakes have a high potential of occurance, and which is constructed in soil

layers with ground conditions corresponding to group 2 in Table 4.5.

(8) Seismic Zone Factor: A factor to decrease the horizontal design seismic coefficient in accordance

with the geographical location of the structural site. The factor has the value of 1.0 for those zones where severe earthquakes have frequently occurred or where severe earthquakes have a high

potential of occurance.

(9) Ground Condition Factor: A factor to modify the horizontal design seismic coefficient depending

upon the ground conditions of the structural site. The factor is intended to unify the margin of

VIII-13
safety against earthquake disturbances among structures constructed in various ground conditions,

in consideration of damage previously experienced.

(!0) Importance Factor: A factor to modify the horizontal design seismic coefficient depending on the importance of the structure.

(11) Inertia Force: Product of the weight of the structural body and the design seismic coefficient.

(12) Seismic Force: Any forces such as inertia forces, earth pressures, etc. to which structures are subjected during earthquakes.

(13) Coefficient of Subgrade Reaction: A coefficient given by the following formula, which is also shown in Section 3.1 of the Volume of Spread Foundations, the Specifications for Design of Substructures of Highway Bridges.

(14) Earthquake Response Analysis: An analysis of the dynamic behavior of structures during

earthquakes.

CHAPTER 2 BASIC PRINCIPLES FOR EARTHQUAKE-RESISTANT DESIGN

(1) The earthquake-resistant design for a highway bridge shall provide sufficient stability against seismic disturbances for the structure as a unit and also for all parts thereof, including superstructures,

substructures, and surrounding soils, considering topographical and geological conditions at the site.

(2) The seismic coefficient method basically shall apply to the earthquake-resistant design of relatively

rigid structures.

(3) The modified seismic coefficient method considering structural response shall apply to the earthquake resistant design of relatively fiexible structures which are of long fundamental periods of

vibration, such as bridges with highrise piers.

The earthquake response analysis shall also be adopted for those structures, for which detailed investigations are required.

(4) Particular attention shall be paid to preventing the fall of superstructures from substructures due to the movements during earthquakes.

When the design allowing for partial failures of the whole system or local failures of certain

structural members is required on the basis of economical considerations, special attention to preventing the fall of superstructures shall be paid.

CHAPTER 3 LOADS AND CONDITIONS IN EARTHQUAKE-RESISTANT DESIGN

(1) The following loads shall be taken into account in earthquake-resistant design. The appropriate loads shall be selected from this list on the basis of the location and the type of the structure.

1. Dead Loads

2. Earth Pressures

3. Hydraulic Pressures

4. Buoyancy or Uplifts

5. Effects of Temperature Change

where

K:

p:5:

coefficient of subgrade reaction in kg/cm'

loading pressures in kg/cm'

displacement in cm

3.1 General

VIII-14
6. Effects of Shrinkage due to Humidity in Concrete Structures (Including Creep Effects)

7. Seismic Effects

8. Effects of Friction at Bearing Supports

9. Effects of Consolidation and Settlement of Ground

10. Effects of Movements of Supports

1 1. Other Loads \

(2) Combination of Loads

Design conditions shall be determined considering the fo'low ng four cases. The combination of

various loads listed above shall be decided in accordance with the provisions in individual Specifications according to the type of the bridge considered. Case 1 Maximum stresses will be expected in the members

Case 2 Maximum reactions will be expected in the foundation soils Case 3 Maximum displacements will be expected at the supports Case 4 The structure will become critical conditions due to overturning, sliding, etc.

3.2 Seismic Effects The following seismic forces shall be taken into account to determine seismic effects in design of a bridge structure.

1. Inertia forces due to the dead weight of the structure: The magnitude of the inertia forces shall be

the product of the weight of the structure and the design seismic coefficient. The design seismic

coefficient shall be obtained in accordance with the provisions in Chapter 4 "Design Seismic

Coefficient."

2. Inertia forces due to the superimposed weight: The magnitude of the inertia forces shall be the product of the superimposed weight and the design seismic coefficient. The design seismic

coefficient also shall be obtained in accordance with the provisions in Chapter 4 "Design Seismic

Coefficient."

3. Earth pressures during earthquakes.

4. Hydrodynamic pressures during earthquakes.

3.3 Inertia Forces

(1) Both for the seismic coefficient method and the modified seismic coefficient method, bridge

substructures shall be subjected to the inertia forces of superstructures during earthquakes as

follows (refer to Fig. 3. 1 ):

(a) Horizontal seismic force acting on A/, (left abutment) shall be

f^AL -RaL fAL (3.1)

where ^^^h H'^ (3.2)

In other words.

f^AL fAL

Ha l

- smallest of or

-jl'hWA

(b) Horizontal seismic force acting on P, (Pier 1) shall be

^A (Hbl = 0 in this case) (3.3)

Har + Hbl = largest of ^khWA + Rbl Sbl

or

(3.4)

VIII-15
In equation (3.4), f^L shall satisfy

(3.5)

(c) Horizontal seismic force acting on Ar (right abutment) shall be

Hbr = khWa- -

Movoble Fixed MovoWe

Support (Sal) Suppt)rr(S«d Sjjpport(SBL) "^•^ iTGirdefA (Wa K | - 2*3 Har

HblFixed Suppat (Sbr)

Girder 8 (We)

A"- Pi Ar

Fig. 3. 1 Inertia Forces of Superstructures

(3.6)

where

f/i I : Static friction coefficient at movable support Ai^

,

fsL ' Static friction coefficient at movable support , ^AL - Horizontal seismic force (inertia force or friction force) acting onAi due to Girder A,

HyiR-. Horizontal seismic force (inertia force) acting onP, due to Girder A,

Hbi : Horizontal seismic force (inertia force or friction force) acting on P, due to Girder B, ^BR • Horizontal seismic force (inertia force) acting on/1^ due to Girder B,

fc/, : Horizontal design seismic coefficient (in accordance with the provisions in Chapter 4

"Design Seismic Coefficient": kf, in equation (4.1) for the seismic coefficient method,

or in equation (4.2) for the modified seismic coefficient method, shall be employed),

Ra l

' Reaction at -4/, due to , R-A R : Reaction at P, due to (V^ , Rbl ' Reaction at P, due to Wg,

Rbr' Reaction at /I due to W'b, (V^ : Dead weight of Girder /4 , and

Wb'. Dead weight of Girder B.

(2) Inertia forces shall be assumed to act at the center of gravity of the structure. In designing the substructure, inertia forces exerted from superstructures may be assumed to act at the level of the base of the supports in the longitudinal direction to the bridge axis, and at the level of the center of

gravity of the superstructures in the transverse direction. In the transverse direction the level of the center of gravity of the superstructures may generally be taken as the lower level of the floor slab.

(3) Inertia forces shall be assumed to come from two horizontal directions: longitudinal and transverse to the bridge axis, or parallel and perpendicular to the principal axis of any structural elements of

the bridge.

3.4 Earth Pressures during Earthquakes

Earth pressures during earthquakes shall be determined in accordance with provisions in the

Specifications for Design of Substructures of Highway Bridges (See Reference 1 at the end of the English

VIII-16
Version). The horizontal seismic coefficient in the calculation of earth pressures shall be in accordance

with the provisions in Section 4.2 "Design Seismic Coefficient in the Seismic Coefficient Method."

3.5 Hydrodynamic Pressures during Earthquakes

Hydrodynamic pressures during earthquakes shall be determined by the following formulas. The

pressures shall be assumed to act in the same direction as that of the inertia forces given by the

provisions in Section 3.3 "Inertia Forces."

{ 1) Hydrodynamic Pressures on Walls Hydrodynamic pressures acting on one side of a wall-type structure shall be determined as follows

(refer to Fig. 3.2):

P =^khWobh' . (3.7)

Ay =y/i - - <3.8)

Fig. 3.2 Hydrodynomic Pressures on Woll Struchwes

Mean

r, Woler Level

S

(orgiojnd surloce)

Average Width

Fig. 3.3 Hydrodynamic Pressures on Coluinn Strudures

where

b: Width of the wall in meters in the perpendicular direction to that of the pressure,

h: Depth of water in meters,

hg-. Height of the total hydrodynamic pressure in meters above the bottom of the water.

VIII-17
kf,: Horizontal design seismic coefficient given by the provisions in Section 4.2 "Design

Seismic Coefficient in the Seismic Coefficient Method,"

P: Total hydrodynamic pressure in t, and

Wc,: Unit weight of water in t/m'

.

(2) Hydrodynamic Pressures on Columns

Hydrodynamic pressures acting on a column-type structure surrounding by water shall be

determined as follows (refer to Fig. 3.3):

P =-jkhWob'h (\ --^) fofy^2 (3.9)

P=i'^hi^ob^h for-^>2 (3.10) o n

fig = (3.11)

3.6 Ground Surface Assumed in Earthquake Resistant Design

In earthquake resistant design the bearing capacities of soil layers which are specified in Section 3.7

"Soil Layers Whose Bearing Capacities are Neglected in Earthquake Resistant Design" shall be neglected.

Ground surface in design shall be assumed to be the level of the lower boundary of the neglected layer, if

the layer extends continuously below the actual ground surface. For this case the neglected layer shall be assumed to have properties of zero cohesion and zero angle

of internal friction. 3.7 Soil Layers Whose Bearing Capacities are Neglected in Earthquake Resistant Design

(1) Sandy Soil Layers Vulnerable to Liquefaction Saturated sandy soil layers which are within 10 meters below the actual ground surface, have a standard penetration test N-value less than 10, have a coefficient of uniformity less than 6, and also have a Dao-value on the grain size accumulation curve between 0.04 mm and 0.5 mm, shall have a high potential for liquefaction during earthquakes. Bearing capacities of these layers shall be

neglected in design. Saturated sandy soil layers which have a Djo-value between 0.004 and 0.04 mm or between 0.5 mm

and 1.2 mm may liquefy during earthquakes, and shall be given particular attention. Estimation

whether or not these layers will liquefy shall be made in accordance with the available information on liquefaction problems.

When a special investigation is performed, the provisions in this item (1) in Section 3.7 may not be

required to apply.

(2) Cohesived Soil Layers and Silty Soil Layers

Bearing capacities of cohesive soil layers and silty soil layers, which are within 3 meters of the actual ground surface, and are very soft such as those with the compression strength, determined by

unconfined compression tests or field tests, less than 0.2 kg/cm^, shall be neglected in design.

(3) Weight of Soil Layers Whose Bearing Capacities are Neglected

The weight of soil layers whose bearing capacities are neglected shall have surcharge effects on the lower ground.

3.8 Buoyancy o'r Uplifts

Buoyancy or uplifts shall be determined in accordance with the provisions in the Specifications for Design of Substructures of Highway Bridges (See Reference 2 at the end of the English Version).

VIII-18
CHAPTER 4 DESIGN SEISMIC COEFFICIENT

4.1 General

The design seismic coefficient shall generally be determined in accordance with the provisions in Section 4.2 "Design Seismic Coefficient in the Seismic Coefficient Method." For those bridges, however,

which have flexible piers and long fundamental periods, such as those with piers higher than 25 meters above the ground surface, the design seismic coefficient shall be determined in accordance with the provisions in Section 4.4 "Design Seismic Coefficient in the Modified Seismic Coefficient Method

Considering Structural Response."

4.2 Design Seismic Coefficient in the Seismic Coefficient Method

(1) The horizontal design seismic coefficient shall be determined by the following formula: kh = Vx .Vi -ko - -(4.1)

where

ki,: Horizontal design seismic coefficient,

ko'. The standard horizontal design seismic coefficient (= 0.2),

fi : Seismic zone factor, V2 ' Ground condition factor, and

i>3 : Importance factor. The value of k;, shall be rounded to two decimals. The minimum value of k/, shall be considered as 0.10. The values of factors Ui , i>2 and shall be obtained by the provisions in Section 4.3 "Factors

for Modifying the Standard Horizontal Design Coefficient."

(2) The vertical design seismic coefficient, k^ may generally be considered as 0. Vertical seismic forces

for design of bearing supports, however, shall be determined in accordance with the provisions in Section 5.3 "Vertical Seismic Forces for Design of Connections between Superstructures and

Substructures."

(3) The horizontal design seismic coefficient may be considered as 0 for structural parts, soils and water below the assumed ground surface in design. The assumed ground surface in design shall be

determined in accordance with the provisions in Section 3.6 "Ground Surface Assumed in Earthquake Resistant Design." Item (3), however, shall not apply to underground structures such as culverts.

4.3 Factors for Modifying the Standard Horizontal Design Seismic Coefficient

(1) Seismic Zone Factor

Seismic zone factor shall be determined in accordance with Table 4.1, in which the zone

classification shall be determined from Fig. 4.1 or Table 4.2.

Table 4. 1 Seismic Zone Factor i")

Zone Value of v,

A 1.00

B 0.85

C 0.70

VIII-19
Fig. 4. 1 Seismic Zoning Map

VIII-20
(Note)

Table 4.2, which indicates in detail the seismic zone through the use of place names, is omitted

in the English version.

(2) Ground Co^idition Factor Ground condition factor shall be determined in accordance with Table 4.3.

Table 4.3 Ground Condition Factor

Group Definitions') Value of 1^2

1

( 1 ) Ground of the Tertiary era or older (defined as bedrock hereafter)

(2) Diluvial layer^) with depth less than

10 meters above bedrock

0.9

2

(1) Diluvial layer^) with depth greater than

10 meters above bedrock

(2) Alluvial layer^) with depth less than

10 meters above bedrock

1.0

3

Alluvial layer3) with depth less than 25 meters, which has soft layer^) with depth less than

5 meters

1.1

4 Other than the above 1.2

(Notes)

1) Since these definitions are not very comprehensive, the classification of ground condition*

shall be made with adequate consideration of the bridge site. Depth of layer indicated here shall be measured from the actual ground surface.

2) Diluvial layer implies a dense alluvial layer such as a dense sandy layer, gravel layer, or

cobble layer.

3) Alluvial layer implies a new sedimentary layer made by a landslide.

4) Soft layer is defined in Section 3.7 "Soil Layer Whose Bearing Capacities are Neglected in

Earthquake Resistant Design."

(3) Importance Factor Importance factor shall be determined in accordance with Table 4.4.

VIII-21
Table 4.4 Importance Factor i'3

Group Definitions Value of ^3

1

Bridges on expressways (limited-access

highways), general national highways and

principal prefectural highways.

Important Bridges on general prefectural

highways and municipal highways.

1.0

2 Other than the above 0.8

Note: The value of may be increased up to 1.25 for special cases in Group 1.

4.4 Desien Seismic Coefficient in the Modified Seismic Coefficient Method Considering Structural Response

The design seismic coefficient specified in this Section shall apply to the design of superstructures

and substructures of those bridges which have flexible piers and relatively long fundamental periods,

such as ones in which the height of the substructures is 25 meters or more above the ground surface in design.

The above-mentioned ground surface in design shall be specified in Section 3.6 "Ground Surface Assumed in Earthquake Resistant Design."

The Design Seismic coefficients in this Section shall be determined by modifying the horizontal design seismic coefficient in Section 4.2 "Design Seismic Coefficient in the Seismic Coefficient Method," on the basis of characteristics of strong earthquake ground motions recorded, and dynamic properties of

bridge structures, such as natural periods, mode shapes and damping capacities of bridge substructures. After obtaining the design seismic coefficient, the method for applying seismic loads for the design

shall be the same as specified by the provisions in Chapter 3 "Loads and Conditions in Earthquake

Resistant Design."

The provisions in thes Section shall not apply to the design of superstructures of those bridges in which superstructures are flexible and have longer periods, such as suspension bridges.

4.4. 1 Design Seismic Coefficient

(1) The horizontal design seismic coefficient shall be determined by the following formula:

>(hm = /Jfc/i— — —- (4.2)

where

kf,„ : Horizontal design seismic coefficient in the modified seismic coefficient method

considering structural response,

kf, : Horizontal design seismic coefficient given by eq. (4.1), and

j3 : A factor dependent on the fundamental period of the bridge, and obtained by

Table 4.5 or Fig. 4.2. For structures whose fundamental periods are shorter than 0.5 sec, /? may be

considered as 1 .0 The value of k;,„ shall be rounded to two decimals. The minimum value of khm shall be considered

as 0.05.

(2) The vertical design seismic coefficient shall be provided in accordance with the provisions in Item

(2) of Section 4.2. (3) The horizontal design seismic coefficient for the portions below the assumed ground surface in

design, shall be provided in accordance \yith the provisions in Item (3) of Section 4.2.

VIII-22
0.301 1 1

\—1—1 1 i 1 1 I I l-J OS lO ZO ZD ^0 iO

Natural Ptriadt T (sec)

Fig. 4. 2 Factor fi (For reference of Table 4. 5)

Table 4.5 Value of (J

Group on Ground

Conditions

Value of /3 for Fundamental Period T (sec.)

1 /?= 1.25

for 0.5 ^T^ 1.1 ^= 1.40/T

for 1.1 ^T^ 2.8

(3= 0.50

for T ^ 2.8

2 /3=1.25

for 0.5 ^T^ 1.4

(3= 1.75/T

for 1.4^T^3.5

(3 = 0.50

for T^ 3.5

3

/3= 1.25

for 0.5 ^T^ 1.7

/J= 2.10/T

for 1.7 ^T ^4.2

(3 = 0.50

for T > 4.2

4

/3= 1.25

for 0.5 ^T^ 2.0

^= 2.50/T

for2.0^T<5.0

^=0.50

for T^ 5.0

(Notes)

1) Refer to Fig. 4.2.

2) Refer to Table 4.3 regarding groups on ground conditions.

4.4.2 Method for Obtaining Fundamental Periods Fundamental natural periods of a bridge shall be determined for the individual system consisting of

each substructure and the part of superstructures supported by it rather than for the structural system as

a whole.

(1) Bridges Supported by Spread Foundations or Pile Foundations

For those bridges which are supported by spread foundations or pile foundations, the foundamental

periods may be obtained from Table 4.6. Any formulas in Table 4.6 shall apply to bridges in which the level of the base of the footing is

lower than that of the assumed ground surface in design and the deformation of the substructure is

mainly caused by the elastic flexural deformation of the pier which is the upper part of the

substructures above the top of the footing. Therefore, they shall not apply to bridges in which the

level of the base of the footing is higher than that of the assumed ground surface in design.

VIII-23
Table 4.6 Fundamental Periods of Bridges Supported by

Spread Foundations or Pile Foundations

Type of Structural System

Direc- tion

Formulas for Fundamental Periods

Material of Pier

Reinforced Concrete Steel

A bridge where most superstructures are continuous,

have fixed supports (or movable supports specified in

I Article 5.2. 1) on most substructures, and also have rigid

abutments, to one of which

the extreme end of the superstructures is connected with a

fixed support (See Fig. 4.3)

Transverse

lo.SWp + IVu , 3

(4.3)

V 4.5EI,

(4.4)

Longitudinal

(4.5)

Other than the above:

2 For example, a bridge with

simple supports

Longitudinal or

Transverse

/0.3Y^^«,3 V 3EIg

(4.6)

p. p.

F Filed Support M McnoUe Suppoft

H Hinqc Supoor'

Fig. 4. 3 An Example of Type 1 of Structural System in Table 4. 6

where

T

K

E:

I:

h : g

Fundamental period in second of the system consisting of a substructure and the section of the superstructures which it supports,

The weight of the pier in t. The weight of the section of superstructures in t supported by the substructure being

considered. Young's modulus of the pier in t/m' , Moment of inertia of the pier in m" in the direction considered. For piers with varying section with the height, I may be an average value. The height of the pier in m, and

Acceleration of gravity ( = 9.8 m/sec^ ).

VIII-24
(Note)

Eq. (4.4) shall apply to those bridges which have a ratio of the length between supports on both

abutments to the width between outside girders, less than approximately 50 (refer to 2/b in Fig.

4.3).

(2) Bridges Supported by Caisson Foundations For those bridges which are supported by caisson foundations, the fundamental period!) may be

obtained from Table 4.7.

Table 4.7 Fundamental Periods of Bridges Supported by Caisson Foundations

Type of Structural System Direction Formulas for Fundamental Periods

1 Type 1 in Table 4.6

Transverse one of eqs. (4.3) (or (4.4), and (4.7), which gives the largest value of ^

Longitudinal eq.(4.5)

2 Type 2 in Table 4.6 Transverse one of eqs. (4.6) and (4.7), which gives

the largest value of (i Longitudinal

(Note)

2if -(4.7)

where

T : Fundamental period in second of the system consisting of a substructure and the section

of the superstructures which it supports,

Wp -. The weight of the pier in t,

Wu'. The weight of the part of superstructures in t supported by the substructure being

considered,

Wc'. The weight of the caisson foundation in t, h : The height of pier in m, A : Cross-sectional area in m^ at the base of the caisson foundation,

Ig Moment of inertia in m" at the base of the caisson foundation in the direction considered,

Kh: Horizontal coefficient of subgrade reaction in t/m' at the level of the base of the caisson

foundation,

Ky : Vertical coefficient of subgrade reaction in t/m' at the base of the caisson foundation,

Ks : Horizontal coefficient of subgrade reaction in t/m^ for shear deformation at the base of

the caisson foundation, and

g : Acceleration of gravity ( = 9.8 m/sec^ ).

VIII-25
CHAPTER 5 GENERAL PROVISIONS FOR DESIGN OF STRUCTURAL DETAILS

S.lGeneral

Every bridge structure or every portion thereof shall be designed and constructed to resist seismic

forces as provided in Chapter 1 through Chapter 4 and to meet the provisions for design of structural

details specified in this Chapter.

Moreover, attention shall be paid to the following respects.

(1) For those abutments which are constructed in soft ground layers, the failure of the ground layer

during earthquakes shall be checked.

(2) For those bridges in which any adjacent substructures have different ground conditions, different type of structural systems, or different structural dimensions, special attention shall be paid to the design of structural details, considering that those two substructures may respond differently during

earthquakes.

(3) For those portions such as joints between superstructures and substructures, connections between

piers and foundations, or connections between footings and piles in pile foundations, where the seismic forces may not be transmitted smoothly and seismic failure have been observed often in the

past, particular attention shall be paid to the design of structural details, considering accuracy of

evaluation of ground conditions, existance of construction joints, accuracy of construction, etc.

5.2 Devices for Preventing Superstructure from FaUing

Movable supports shall have stoppers (special devices for resisting large movements of superstructures during earthquakes) to prevent the superstructures from falling from the substructures, caused by the dislocation of the upper shoes of the supports from the lower shoes during strong

earthquakes (refer to 5.2.1). For the girder ends one of the following methods shall be employed as well as the above-mentioned

consideration.

(1) A method extending the length between the end of the support and the edge of the substructure (or

widening the width of the crest of the substructure in the longitudinal direction to the bridge axis)

in order to prevent the superstructures from falling from the substructure (refer to 5.2.2 or 5.2.3).

(2) A method connecting adjacent girders on the substructure to prevent the superstructures from

falling from the substructure even if they become dislodged from the substructure (refer to 5.2.4).

5.2.1 Stoppers at Movable Supports

The allowable movable length in the design of stoppers at movable supports shall be assumed as the sum of the movement due to temperature change, the movement due to the deflection of the girder

when subjected to live loads, a margin for covering construction errors, and 20 mm.

The above-mentioned provision need not apply to those stoppers which are not installed near the

supports.

The horizontal design seismic coefficients for designing stoppers shall be determined by increasing the horizontal design seismic coefficient given by eq. (4.1) or (4.2) by 50% or more.

5.2.2 Method of Extending the Length between the End of the Support and the Edge of the Substructure For those substructures which support the ends of girder, the length 5(in cm) between the end of

the support and the edge of the substructure, shall be equal to or more than the value given by the following formulas:

5 = 20 + 0.5 / for I <. 100 m .

5 = 30 + 0.4 / for / > 100 m

VIII-26
where

S : Length between the end of the support and the edge of the substructure in cm, and

/ : Span length in meters. For particularly important bridges constructed in soft ground layers (Group 4 in Table 4.3), the

value of S shall be equal to 35 cm or more.

5.2.3 Suspended Joints For suspended joints the length between the ends of girders shall be equal to 60 cm or more, as shown in Fig. 5.1. The length for those bridges constructed in soft ground layer (Group 4 in Table 4.3)

shall be equal to 70 cm or more.

5.2.4 Method of Connecting Adjacent Girders Devices for connecting adjacent girders on substructures shall have the movable length specified in Section 5.2.1, for cases in which at least one of two supports is a movable one. The devices shall be designed to rotate freely to allow the rotation of the girder subjected to live

loads, for cases in which both two supports are fixed on one pier.

5.3 Vertical Seismic Forces for Design of Connections between Superstructures and Substructures The vertical design seismic coefficient for the design of connections between superstructures and

substructures shall be assumed as 0.10. When the vertical design seismic coefficient applies upward, only

seismic forces shall be considered, neglecting the effects of the dead loads. The same value of the vertical design seismic coefficient shall be employed for the design of any

connections similar to the above.

5.4 Methods for Transmitting Seismic Forces at Connections between Superstructures and Substructures The method for transmitting seismic forces at connection between superstructures and substructures

shall be as follows:

(1) For cast-in-place reinforced concrete bridges, the means of transmission of seismic forces shall be due to the bearing pressure between the swelling at the base of the lower shoe and the concrete at

the crest of the substructure. The concrete portion near the base of the lower shoe shall resist

seismic forces as one body together with the pier of the substructure. In the above-mentioned cases,

the means of transmission of seismic forces between the upper shoe and the girder shall be due to

the anchors fixed on the upper shoe. For a margin of safety, anchor bolts between the lower shoe and the substructure shall be designed

to resist seismic forces alone, in consideration of cases where no resistance between the swelling at

the base of the lower shoe and the concrete at the crest of the substructure can be expected.

The above method is recommended not only for cast-in-place concrete bridges, but, if possible, also

for prefabricated concrete bridges or steel bridges.

60°" or mare

Fig. 5. 1 Length between the Ends of both Girders

VIII-27
(2) In cases where no bearing resistance of concrete can be expected, anchor bolts shall be employed to transmit the seismic forces. In these cases one of the following two methods shall be considered.

(a) A method in which a steel plate with anchor bolts is fixed firmly on the crest of a substructure while concrete is being placed, and then the lower shoe of the support is welded to the steel plate after the erection of the girders.

(b) A method in which a hole is prepared when concrete is placed, a support is set up near the hole,

and then anchor bolts are fixed by placing cement mortar into the hole, or a method in which

anchor bolts are set up while concrete is being placed, and then the lower shoe is fixed on the anchor bolts.

(3) Anchor bolts used shall be 25 mm or more in diameter, and the depth of the anchor bolts fixed in the concrete shall be 10 times the diameter or more.

5.5 Devices Expected for Decreasing Seismic Forces When any devices which are expected to decrease seismic force are employed for bridee structures,

sufficient investigations shall be conducted on their effectiveness, and special attention shall be paid to preventing the superstructures from falling.

CHAPTER 6 MISCELLANEOUS PROVISIONS

When sufficient reasons exist, these Specifications need not apply to the design of bridges.

REFERENCE

References are provided exclusively in the English version. (Reference 1] Specifications for Design of Substructures of Highway Bridges

Volume for General Survey and Design

Part 3 Design, Chapter 2 Loads. Section 2.5 Earth Pressures Earth pressures acting on a wall shall be the distributed loads given by the following formulas:

( I ) Normal Earth Pressures

(a) Earth pressures acting on a movable wall during normal time shall be determined by the Coulomb's theory as follows:

i) For Sandy Soils

Pa = y-KyiX + KAq

Pp = y-Kpx+Kpq

ii) For Cohesive Soils

Pa = y Kj^x - ICsTK^ + KaQ

Pp = yKpx + ICyJTp + KaQ

(b) Earth pressures on a fixed wall during normal time shall be determined by

VIII-28
(2) Earth pressures during earthquakes shall be determined by the Mononobe-Okabe method.

Pa = ( \ -ky) y x KEA

Pp = (.1 - ky) y X Kep

Active Eofth Pressure

Possive Eorttt Pressure

Fig. 1 Earth Pressure

where

C ' Cohesion of the soU in t/m^

,

Kji ' Active earth pressure coefficient for Coulomb's theory,

Kp ' Passive earth pressure coefficient for Coulomb's theory, f^EA : Active earth pressure coefficient during earthquakes,

Kgp: Passive earth pressure coefficient during earthquakes,

Kg : Earth pressure coefficient at rest,

fc„ : Vertical seismic coefficient,

Pji : Active earth pressure in t/m^ at depth of x meters, Pp ' Passive earth pressure in t/m^ at depth of x meters, q : Surcharge in t/m^ on ground surface, X : Arbitary depth in meters, a : Angle between the ground surface line and the horizontal line, 7 : Unit weight of the soil in t/m'',

5 : Angle of friction between the wall and the soil, 0 : Angle of internal friction of the soil, and

6 ' Angle between the back line of the wall and the vertical line.

VIII-29
(Comments)

K/i , Kp, Kea and l^EP are expressed as follows:

COS^ (0-6)

cos^e COS (5+6) 1 +

sin (0 + 6) • sin (0 - a)

COS (d + 6) cos(0 - a)

Kp

cos^ (0 + 6)

cos^d • cos(0 +6)

KEA

Kep •

/ sin (0 - 6) • sin (0 + a)

^ V cos(0 +6).. cos (5 -a)_

cos^ (0 - 6o - g)

cos 00 cos^e • cos(fl +0o) 1 +

sin 0 sin (0 - g - 00

)

cos(0 -eo) cos(0 -Q)

cos^ (0 - 00 + 5)

cos 00 • COS^0 • COS (0 — 0o)

/sin 0 sin (0 + a — 0o)

~4 cos(0 -0o) -cos(0 - a)

where sin (0 ± a - 0o) = 0 when 0±a-0o<O

and 5 is assumed to be zero during earthquakes.

where

= tan

, kh

1 -fc„

fc/,: Horizontal seismic coefficient

ky-. Vertical seismic coefficient

[Reference 2)

Specifications for Design of Substructures of Highway Bridges

Volume for General Survey and Design

Part 3 Design, Chapter 2 Loads

Section 2.7 Buoyancy or Uplifts

When it is apparent that buoyancy forces or uplifts act on structures, they shall be taken into account in the design.

(Comments)

When it is unknown whether they act or not, both cases shall be taken into account in the

design.

VIII-30
ON SPECIFICATIONS FOR EARTHQUAKE-RESISTANT DESIGN

OF THE HONSHU- SHIKOKU BRIDGES (JSCE 1974)

by

Ishio Kawasaki

Director, Honshu-Shikoku Bridge Authority

and

Eiichi Kurbayashi

Chief, Earthquake Engineering Research Section

Public Works Research Institute

Ministry of Construction

ABSTRACT

The following describes the general specifications for earthquake resistant design

of the Honshu-Shikoku Bridges, as developed by JSCE in 1974, after significant study.

Key Words: Bridges; Earthquake Design; Earthquake Forces; Specifications; Seismic Provisions.

VIII-31
Introduction

The Japan Society of Civil Engineers issued "Specifications for Earthquake-Resistant

Design of the Honshu-Shikoku Bridges" ^'^^ in May, 1967. The Specifications were developed

by JSCE's Sub-Committee (Chairman: Professor Shunzo Okamoto) , which existed fora

five year period. The committee stared in 1962 with a commission jointly from the

Ministry of Construction and the Japanese National Railways (the Japan Railways Construction Public Corporation took over the work of JNR after March, 1964) . In May, 1970 three

years after the completion of the above Specifications (1967) , the Earthquake Engineering

Committee (Standing Committee) of the JSCE established a Joint Meeting for Studying Earthquake-Resistant Design of the Honshu-Shikoku Bridges (Chairman: Professor Shunzo Okamoto)

,

and to consider future research needs. The Joint Meeting was organized in cooperation

with the regular members of the Earthquake Engineering Committee, the staff members of the

Japan Highway Public Corporation, the Japan Railways Construction Public Corporation and

the Honshu Shikoku Bridge Authority (the Authority was established in July, 1970) , and

experts from various organizations. The committee attempted to improve the Specifications

(2) (1957) in view of recent progress in the area, and presented in June, 1971 a report

summarizing the results of its research activity.

Furthermore, with a commission from the Honshu-Shikoku Bridge Authority, the JSCE

established a Research Sub-Committee on Earthquake-Resistant Design of the Honshu-Shikoku

Bridges (Chairman: Professor Keizaburo Kubo) in June, 1971. The Research Sub-Committees

charge was to amend the Specifications for Earthquake-Resistant Design of the HonshuShikoku Bridges (1967) and also to clarify the design procedure details. The Research

S\ib-Committee studies showed that the following three subjects should be investigated

extensively;

1) Evaluation of seismic forces —Effects of near earthquakes on structures, effects

of long-period ground motions and their measuring systems, factors to be considered in determining the magnitude of seismic forces, etc.

2) Evaluation of dynamic characteristics of soils and foundations —Dynamic characteristics of multi-column foundations and caisson foundations with an emphasis in

obtaining a design procedure based on dynamic analysis of foundations, investigations on earthquake-resistant design practices for foundations, etc.

3) Earthquake-resistant design practices for bridges with span length of 200 m or more (such as suspension bridges, truss bridges, etc).

The Research Sub-Committee concluded investigations on those subjects by March, 1974,

(3) and recently published the final report . The report proposes revised Specifications

for Earthquake-Resistant Design of the Honshu-Shikoku Bridges (1974) . Although the new Specifications (1974) include the commentary which gives additional explanations necessitated for design practices, the main body of the Specifications will be introduced herein.

VIII-32
Specifications for Earthquake-Resistant Design of the Honshu-Shikoku Bridges (JSCE-1974)

1. General

1 . 1 Scope

The provisions in the Specifications apply to earthquake-resistant design of the

Honshu-Shikoku bridges.

1.2 Notations

The following notations are used in the Specifications:

Notation

A

a

d

h

%

X

a

Y,w

Definition

Lateral cross-sectional area of a structure

Length of cross-section of a structure in

the parallel direction to that of seismic

motion considered

Width of cross-section of a structure in

the perpendicular direction to that of

seismic motion considered

Depth of water

Damping ratio

Design seismic coefficient

Response seismic coefficient

Response acceleration spectrum

Response velocity spectr\im

Natural period of a structure

Any depth below the surface of water

or ground

Coefficient dependent on vibrational

modes of a structure

Coefficient dependent on shape of a foundation

Unit weight of sea water

Modification factors necessitated in

obtaining design seismic coefficient from

response seismic coefficient

Unit

2

m

m

m

m

cm/sec

cm/sec

sec

m

t/m"

VIII-33
2. Earthquake to be Considered in Design

2.1 Earthquake to be Considered in Design

An earthquake with the following characteristics shall be considered in design:

1) Magnitude: Large-scale (Namely around 8 on the Richter scale)

2) Location: Comparatively far from the bridge sites (Namely off Kii Peninsula,

or off Tosa)

3) Frequency: Once or twice per one hundred years

2.2 Design Ground Acceleration

The maxim\am value of the design ground acceleration shall be 180 gals at the level

of the surface of ground layers which support foundations at the bridge sites.

3. Basic Principle for Earthquake-Resistant Design

3.1 Method of Design Calculation

1) For structures whose principal dimensions are determined by some requirements

other than those by earthquake resistant design, the modified seismic coefficient method considering structural response shall be adopted. The results

shall generally be examined through a dynamic analysis.

2) For structures whose principal dimensions can be determined by requirements

from earthquake-resistant design, the structural dimensions shall be determined

by the response spectrum method of dynamic analysis. In such cases the results

should be examined through a numerical integration method of a dynamic analysis

4. Design by the Modified Seismic Coefficient Method Considering Structural Response

4.1 Design Procedure

In a design, based on the modified seismic coefficient method when considering

structural response, the seismic forces specified in 4.2 and additional effects

specified in 4.4 shall be taken into account simultaneously.

4.2 Seismic Forces

Seismic forces shall be determined by the product of structural dead weight and

the design seismic coefficient. The design seismic coefficient is provided in "4.3

Design Seismic Coefficient Considering Structural Response."

4.3 Design Seismic Coefficient Considering Structural Response

4.3.1 Horizontal Response Seismic Coefficient

The horizontal response seismic coefficient (K ) shall be determined from Fig. 1.

R

This figure was obtained by assuming the following conditions:

1) Foundations are constructed directly on the Tertiary layer (or older) at

the site where the surface Quaternary layer is shallow or none.

2) The maximum ground acceleration for design is expected to be 180 gals.

4.3.2 Vertical Response Seismic Coefficient

The Vertical response seismic coefficient shall be generally the half of the

horizontal response seismic coefficient.

4.3.3 Design Seismic Coefficient Considering Structural Response

The design seismic coefficient, considering structural response, shall be determined by the following formula:

VIII-34
D 1 2 3 4 5 R

where K^^: Design seismic coefficient considering structural response

K^: Response seismic coefficient (see Fig. 1) dependent on structural

type and the predominant natural period.

y : Modification factor dependent on the maximum ground acceleration for

^ design {y^=1.0 for the case of 180 gals)

y : Modification factor to cover the influences of the higher modes

2 y : Modification factor to apply equivalent uniform seismic loads

3 y : Modification factor to cover the case where the direction of structural

response is perpendicular to that of seismic motion applied.

y^ Modification factor to be adjusted by engineering judgment.

The predominant natural period herein is the natural period corresponding to a vibration mode whose response (stress or displacement) is the most predominant in

respective structural members. This predominant natural period does not necessarily coincide with the fundamental period.

4.4 Additional Effects to be Considered in Design

4.4.1 Effects of Ground Reactions on the Structural Response and Stresses

In evaluating the effects of the surrounding soil on the structural response

and stresses, ground reactions shall be taken into account as the product of

the structural displacement relative to the ground and the spring constant of

the ground.

4.4.2 Effects of Surrounding Soil and Water on Structural Response

1) The surrounding soil will effect the structural response. Therefore the

spring and damping factors shall be taken into account. The mass effect of

the soil, however, may generally be neglected.

2) The effects of water shall be taken into account by applying the virtual or

added mass, as described in "5.5 Effects of Surrounding Water on Structural

Response".

5. Dynamic Analysis

5.1 Methods of Analysis

The following methods shall be employed in the dynamic analysis to obtain the

structural response.

1) Response spectrum method: Calculate the maximum values of the structural

earthquake response based on the response acceleration spectra specified in 5.3.

2) Numerical integration method: Calculate the time history of the structural

earthquake response based on specific seismic motions specified in 5.4.

5.2 Application of Dynamic Analysis

In performing the earthquake-resistant design of structures, the dynamic analysis

shall be employed in the following ways:

1) The design results obtained by using the modified seismic coefficient method

considering structural response or the seismic coefficient method are

examined by the dynamic analysis.

VIII-35
2) The response spectrum method of dynamic analysis is employed for structures

whose dimension can be assumed by normal loads. The structural dimensions

determined by this procedure are then examined by the numerical integration

method of dynamic analysis.

5.3 Response Acceleration Spectriam for Dynamic Analysis

The response acceleration spectral curves shown in Pig. 2 shall apply to the

response spectrum method of dynamic analysis specified in 5.1^^^

5.4 Seismic Ground Motion for Dynamic Analysis

Seismic ground motions used for the numerical integration method of dynamic anal'

(2) ysis specified in 5.1 shall be either of the following. The maximum acceleration

shall be adjusted to 180 gals by proportioning the various original acceleration records .

1) Typical seismic records obtained near the bridge sites

2) Strong motion records obtained at El Centre in 1940.

5.5 Effects of the Mass of the Surrounding Soil on the Structural Response

In the dynamic analysis of substructures, the effects of the mass of the soil

on the response may generally neglected.

5.6 Effects of Surrounding Water on the Structural Response

For the portion of the structure which is in water, the virtual mass of the

water converted from the hydrodynamic pressure shall be considered by the following

formula

:

M

wx

= a3

^ a/T~

g d

where

:

M : Virtual mass (or added mass) per unit width at the depth of x below the

water surface (t sec2/m2)

a: Coefficient dependent on vibrational mode of the structure

3: Coefficient dependent on shape of the foundation

1) 3 = -(1--^) for the case of ^ < 2 a 4d d —

2) 3 = - (0.7 - T^) for the case of 2 < ^ < 4 a lOd = d =

i 3 yw: Unit weight of the sea water (t/m )

A: Lateral cross-sectional area of the foundation

2 g: Gravity of acceleration (=9.8 m/sec )

a: Length of cross-section of the foundation in the parallel direction to

that of the seismic motion considered (m)

b: Width of cross-section of the foundation in the perpendicular direction

to that of the seismic motion considered (m)

d: Depth of water at the site (m)

VIII-36
5.7 Dynamic Characteristics of Structures

1) The Directions of Seismic Motion to be Considered are;

Directions of seismic motion to be considered in dynamic analysis shall be

longitudinal, transverse, and vertical.

2) Natural Frequencies and Mode Shapes:

The structural earthquake response shall be analyzed by taking into account

the order of the natural frequencies and the corresponding mode shapes. These

conditions are necessary in order to obtain the precise maximum response in

the special consideration for erection and completed construction.

3) Damping ratios:

The damping ratios, used for dynamic analysis, shall be determined in view

of the results of the appropriate investigations.

6. Safety Considerations in Earthquake-Resistant Design

6.1 Factors of Safety in the Modified Seismic Coefficient Method Considering

Structural Response

6.6.1 Combination of Loads

For superstructure: Dead Load + live Load during earthquake + effects of temperature change + seismic effects + effects of movements of

supports + effects of erection errors.

For substructure: Loads from superstructure + deal load + soil pressure +

water pressure + buoyancy or uplift + seismic effects

6.1.2 Increase in Allowable Stresses for Steel Superstructure

The Increase in the allowable stresses for steel superstructures in earthquake

resistant design shall be as follows:

For suspension bridges and long-span bridge: 1.5

For bridges other than the above: 1.7

6.1.3 Stability of Substructures

(1) Allowable bearing capacities of soil: The ultimate bearing capacity of soil

shall be evaluated in accordance with the Specifications for Design of

Substructures of Highway Bridges-Volume for Design of Spread Foundations

(issued by the Japan Road Association) . The minimum values of the allowable

bearing capacities shall be obtained by dividing the ultimate bearing

capacities by the factors of safety specified in Table 1.

(2) Stability for Overturning

In normal design, the position of the resultant force acting on the foundation base shall be located within the middle third of the base. In earthquake-resistant design the force shall be located within the middle two

thirds of the base. When the position of the resultant force, for the

earthquake-resistant design, is outside of the middle two thirds, stability

and deformation of ground and structures shall be examined.

VIII-37
(3) Stability for Sliding

The sliding resistance at the foundation base shall be evaluated in

accordance with the Specifications for Design of substructures of Highway

Bridges-Volume for Spread Foundation (issued by the Japan Road Association)

.

The fractures safety for sliding shall be provided in Table 2

.

(4) Displacement Standards for Substructures

Displacements of substructures shall be generally less than the displacement

standards provided i Table 3.

.2 Factors of Safety in the Design Based on Dynamic Analysis

Factors of Safety, of the design using the dynamic analysis, shall be in accordance with the provisions in 6.1.

VIII-38
References

1. Japan Society of Civil Engineers, "Technical Report of Investigations on the HonshuShikoku Bridges," July 1967. (in Japanese).

2. Japan Society of Civil Engineers, "Report of Research for Improving the EarthquakeResistant Design Criteria for the Honshu- Shikoku Bridges," June, 1971. (in Japanese)

3. Japan Society of Civil Engineers, "Technical Report of Investigations on EarthquakeResistance of the Honshu- Shikoku Bridges," September, 1974. (in Japanese).

VIII-39
Fig. 1.

qals

500'

AOO

330

300

230

200

§100

ofcI 50

§. 30

VIat ce 20

10

0-1

0.2 0.3 o4 0.5 r 2

Natural Period (T)

Response Seismic Coefficient Used in Design

/osec

h=0.02 h: C)amping Ratio

1 0.05

n f n

0.2

Fig. 2.

0.3 0.4 05 1

Natural Period (T)

A 5 10sec

Response Acceleration Spectral Curves

(for maximum acceleration of 180 gals)

VIII-40
M

Direction of

Seismic Motion

Fig. 3, Virtual mass of water

Table 1. Factors of Safety for Bearing Capacities

Construction Manner of Foundation Base

Dry Under Water

Normal Design :i. 0 4. 5

Earthquake -Resistant

Design 2. 0 3. 0

Table 2. Factors of Safety for Sliding

Construction Manner of Foundation Base

Dry Under Water

Normal Design 2. 0 2. 5

Earthquake -Resistant

Design

1. 2 1. 5

Tnhle 3, Displacement Standards for Substructures

Suspension Bridges Other Bridges

Abutments Tower Foundations

Horizontal displacement at the level of

ground surface (cm)

Horizontal

displacement at

the Saddle Position

(cm)

Rotation at the tower

base

^ = 0. 017^ ^ = 0. 0055^ + 2 d = 1

I : central span length (m)

VIII-41
RECENT REVISION OF DESIGN STANDARDS ON

SEISMIC EFFECTS FOR PORT AND HARBOUR STRUCTURES

by

Satoshi Hayashi

Head of Structures Division

Port and Harbour Research Institute

Hajime Tsuchida

Chief of Earthquake Resistant Structures Laboratory

Structures Division

Port and Harbour Research Institute

Setsuo Noda

Chief of Subaqueous Tunnels and Pipe Lines Laboratory

Structures Division

Port and Harbour Research Institute

ABSTRACT

The design standards for port structures have been compiled four times in Japan.

In these design standards, provisions on earthquake resistant design of wharves are

included.

In 1950 the first design standard was published, and in 1959 and 1967 new design

standards as an expansion of preceding one were compiled. Those design standards were recognized as the most advanced design procedure in the times and used very widely for

design of port structures. However, the standards were not related to any law.

In 1973 Port and Harbour Law was revised and to secure the safety in ports it was assigned to establish engineering requirement of facilities in ports. The requirement has been effective since 1974 and the earthquake resistant design of the facilities

are specified in it.

Key Words: Structures, Ports, Harbors, design, specifications, earthquakes.

VIII-42
1. Introduction

A port is an interconnection between sea transportation and land transportation

and is an essential part in the activity of our modern society. Loss or even partial

damage of a port, due to an earthquake, causes serious ffect on the activities of a com- munity. This is especially true after a destructive earthquake, as the port is required

to function for transportation of emergency goods and materials for reconstruction of dam

aged facilities.

During past earthquakes, however, serious damage to port facilities has been

experienced. For instance, in the Niigata earthquake of June 16, 1969, which had a magni

tude 7.5, the damage to the port facilities in the Niigata port cost 22 billion yen.

Wharves are major facilities in a port, and many wharves are earth retaining structures

such as a gravity type quay wall and a sheetpile bulkhead. These structures support

large soil areas which move in a very complex manner during earthquakes. This is one of

the reasons why the port facilities were greatly affected by past earthquakes.

Because of these circumstances, intensive efforts have been made by port engi- neers to increase the aseismicity of port facilities, with a minimum increase of construe

tion cost, which was acceptable to the community.

In this paper, the basic requirements and procedures for earthquake resistant

design of port facilities will be presented. It is well known that a large variety of

structures surrounding a port exists; however, in this paper, the authors will limit the

discussion to sea walls, piers, etc., and those types of structures which exist only in a

port. Therefore, the term "port structures" will mean such structures. In this report,

the design standard means a compilation of typical procedures and considerations. Such

standards would be allowable stresses and factors of safety for designing port structures

These design standards have been published previously four times in Japan, and their

degrees of restriction to actual designs differ according to each publication.

2. History of Design Standard

The first design standard for the port structure was published by the Japan Port

and Harbour Association in 1950, and was called "Manual of Harbour Construction Work

(Title in Japanese: Kowan Koji Sekkei Shiho Yoran) "

^"''^

. The manual consisted of three

parts; namely the recommendations for design of quay walls and piers, the recommendations

for planning and execution of dredging and fill-up, and the recommendations for design of

breakwaters

.

In the recommendations for design of quay walls and piers, relative to earthquake

resistant design, inertia force, earthpressure and dynamic water pressure were considered

For the estimation of the earthpressure during earthqudces, a formula proposed by Matsuo

based on his experimental study was adopted. However, Mononobe-Okabe ' s formula ^"^^

for earthpressure during earthquakes was also explained as an applicable formula.

The second design standard for the port structure was published by the Japan Port

and Harbour Association in 1959, called "Design Manual for Harbour Construction Work

VIII-43
(Title in Japanese: Kowan Koji SekKei Yoran) . This document has been called the

"Green Book" by port engineers because of the color of the cover.

In the "Green Book" the area of Japan was divided into three regions and a seismic

coefficient was specified for each region. The formula to estimate earthpressure , during

earthquakes, consisted of Mononobe-Okabe ' s formula.

The third design standard was published also by the Japan Port and Harbour

Association in 1967, and was called "Design Manual of Harbour Structures in Japan (Title

in Japanese: Kowan Kozobutsu Sekkei Kijun) . This design standard had various unique

characteristics, in comparison to the previous two design standards. The first quality

was that the third design standard presented procedures for the design of the structures,

such that even a less experienced engineer could design the structures without any other

textbook, and it also presented to the engineers the background information on the

procedures. As a result, the third design standard has become a piablication of many pages

and has the characteristics of a specialized textbook.

The second quality was that the third design standard was to be revised regularly,

in order to introduce results of the most recent research and technological developments.

In order to make such a condition possible and easier, the design standard was compiled

such that any page could be replaced by a newly printed page.

The provisions in the third design standard, regarding earthquake resistant design, were presented at the First Joint Meeting, U.S. -Japan Panel on Wind and Seismic Effects,

( 6

)

UJNR , and at the Second Joint Meeting relative to the comments on the revision of the

(7) of the third design standard

The third design standard described previously was recognized as the best procedure at that time by the port engineers. Most of the port structures were designed

according to these design standards

.

3. Engineering Requirement by Port and Harbour Law

In 1973 the Port and Harbour Law was revised. Because of social demands on the

safety of facilities in ports and harbours, the ministry established engineering requirements on facilities which were to be constructed in ports and harbours. In 1974, these

engineering requirements were established as Ordinance No. 30 by the Ministry of Transport,

which prescribes that the facilities in ports and harbours must be safe against earthquakes

as well as dead load, water pressure, wave force, surcharge, impact and drag due to ships,

etc. However, details on earthquake resistant design, such as design procedures, factors

of safety, and allowable stresses, have not been given in the ordinance, but were

specified in an order by the Director General of Bureau of Port and Harbours. Such

a legal system was chosen because considerable progress in earthquake resistant design

is expected in the near future, and flexibility to revise the requirements is necessary.

Even in the order of the Director of Bureau of Ports and Harbours only important

points are specified. Therefore, it is recommended that the third design standard should

be used as a supplement to this order.

VIII-44
The basic consideration of earthquake resistant design of structures, constructed

mainly in ports, is the seismic coefficient method; however, earthquake resistant design

depending on dynamic analysis is also acceptable.

4. Earthquake Resistant Design in the Requirement

In this section the earthquake resistant design, specified in the engineering re- quirements, will be described. However, the following are not a translation of the provisions, but an explanation of the earthquake resistant design. The provisions are

(8) available in the separate publication .

4.1 Design Earthquake Load

1) Earthquake load

The earthquake loads acting on port structures should be calculated by the following

formulas, where the most severe conditions should be chosen.

i) Earthquake load = Dead load x Seismic coefficient

ii) Earthquake load = (Dead load + Surcharge) x Seismic coefficient

The seismic coefficient is defined in the next paragraph.

2) Seismic coefficient

The seismic coefficient should be calculated by the following formula, taking

regional seismicity, foundation soil and importance of the structure into consideration.

Seismic coefficient = Regional seismic coefficient x Factor for subsoil condition x Importance factor

In general the earthquake load is applied horizontally at the center of gravity of the

structure. The earthquake load in the vertical direction is not considered, with the

exception of those special structures which are influenced by a vertical load. The

seismic coefficient should be calculated to two decimal places, where the last digit is

set equal to 0.1, if equal to or greater than 0.08; if the digit is between 0.07 and

0.03 set the value equal to 0.05.

3) Regional seismic coefficient

Standard values of the regional seismic coefficient are tabulated in Table 1. The

seismic coefficient in a region which is not described in Table 1 is determined considering seismicity of the region and the regional seismic coefficients in the neighboring

regions as given in Table 1.

Fig. 1 shows the regional seismic coefficient and Fig. 2 the expected maximum

acceleration of earthquakes in the next 75 years estimated by Kawasumi.

VIII-45
4) Factor for subsoil condition

The standard value of the factor for subsoil condition should be determined as

shown in Table 2.

The classification of the subsoil condition should be assigned as shovm in Table 3, con- sidering the thickness of the quaternary deposit and the kind of subsoil.

5) Factor depending on importance of structure

The standard value of the factor relative to the importance of the structure

should be determined by Table 4.

4.2 Earthpressure in an Earthquake

Lateral earthpressure of sandy soil in earthquakes is computed by using MononobeOkabe's formula which is derived from Coulomb's formula by statically applying the earthquake load to the soil mass in question. For horizontal ground surface, the formula is

as shown in (Fig. 3)

.

2 where f p : intensity of lateral earthpressure in earthquakes (t/m ) 2 w : intensity of uniform load on the ground surface (t/m )

k : seismic coefficient

(j) : angle of internal friction of sandy soil (°)

for general case ...30°

for particularly good backfill ...40'

3 unit weight of soil (t/m ) ; buoyed unit weight should be used below water

level and the following are the standards:

above water table in backfill ...1.8t/m^

3

below water table in backfill ...l.Ot/m

h : depth from the ground surface (m)

K : coefficient of lateral earthpressure

y : angle between wall surface and the vertical (°)

6 : angle of friction between soil and wall (°) ; usually |6| 15°

9 : angle given by the following equations; 9=tan ^ k or 0=tan ^k'

C : angle between failure surface and horizon (°)

In Eqs. (2) and (3), upper signs are for active and passive earthpressure for

typical values of (}> and 6 are shown in Figs. 4 and 5 respectively.

For backfill, in layers (w + Y h) in Eq. (1), these terms should be replaced by

the vertical effective stress. Below the water table the apparent seismic coefficient

should be used. In calculation of the vertical effective stress the bouyed unit weight

of soil is used, this causes a smaller estimation of the earthquake load acting on the

soil mass, since the weight of soil mass in the air should be multiplied by the seismic

coefficient. The apparent seismic coefficient compensates the difference. The apparent

seismic coefficient is given by the following equation

^ Y'-l

where; k' : apparent seismic coefficient

k' =-yrz^ k (4)

3 Y' : unit weight of saturated soil (t/m )

VIII-46
4.3. Dynamic Water Pressure

Rigid wall- and colimn-like structures facing the water are designed by taking the

dynamic water pressure into consideration. However, the structure is retaining soil and

the earthpressure is calculated by Mononobe-Okabe ' s formula with the apparent seismic

coefficient, where the dynamic water pressure is not considered in the design calculation. The reasons for this assumption are: first, when the apparent seismic coefficient

is applied for the earthpressure calculation the inertia force due to pore-water has

already been considered. Also experience from structural performance during past

earthquakes do not indicate the necessity to consider the dynamic water pressure acting on the face of the earth supporting structures.

4.4. Allowable Stresses

Allowable stresses of the materials are determined as shown in Tables 5 through 9.

Allowable stress of tie rods must be equal to or below 40 percent of the certified

yielding stress.

Allowable stresses of steel and concrete, for short period load such as earthquake

load, may be equal to or smaller than 1.5 times of the allowable stresses in normal

condition.

4.5. Earthquake Resistant Design Based upon Special Study

When the seismic coefficient is determined from consideration of a survey on seismicity of the region, and from the characteristics of earthquake motion and

response characteristics of the ground against earthquakes, the requirement described in

4.1 need not be applied to design.

When the earthquake resistant design is confirmed by consideration of the dynamic

characteristics of the structure and the investigation on the response analysis for the

structure against earthquakes, the requirement described in 4.1 need not be applied.

It is advised that in case of necessity the earthquake resistant design is confirmed by the consideration of the dynamic characteristics of the structure and the

investigation on the response analysis against earthquake.

4.6 Structures Other Than Port Structures

Structures other than port structures are normally designed by the design standards

established for each type of structure.

1) Design Specifications for Steel Highway Bridges, established by the Japan Road Association, may be applied to the earthquake resistant design of the highway bridges.

2) Design Specifications for Steel Railway Bridges by the Japan Society of Civil

Engineers or Design Specifications for Civil Engineering Facilities established by the

Japan Railway Facility Association may be applied to the earthquake resistant design of

the railway bridges.

VIII-47
3) On the earthquake resistant design of the pipe line for transporting such fluid

and gaseous materials as petroleum, Fire Service Law, Oil Pipelines Enterprise Law

or High Pressure Gas Control Law and regulations established to supplement these laws,

may be applied.

4) On the earthquake resistant design of the buildings. Building Standard Law and

regulations established to supplement the law may be applied.

5. Future Follow-up

In this report the earthquake resistant design designated by the Port and Harbour

Law, has been presented. Many new types of structures are going to be constructed

on or off-shore and the social demands for environmental consideration of such structures

have been increasing. Because of such circximstances , revision of the Design Manual of

Harbour Structures in Japan has been undertaken. The rationalization of earthquake

resistant design will be achieved by this revision.

VIII-48
References

1) Manual of Harbour Construction Work, The Japan Port and Harbour Association,

May 1950, 115p.

2) Haruo Matsuo: Experimental Study on Distributions of Earth Pressure Acting on Retaining Walls during Earthquakes, Journal of the Civil Engineering Society,

Vol. 27, No. 2, February 1941, pp. 1-24

3) Saburo Okabe: General Theory on Earthpressure and Seismic Stability of Retaining

Wall and Dam, Journal of the Civil Engineering Society, Vol. 10, No. 6, December

1924, pp. 1277-1323.

4) Design Manual of Harbour Construction Work, The Japan Port and Harbour Association,

June 1959, 453p.

5) Design Manual of Harbour Structures in Japan, The Japan Port and Harbour Association,

April 1967.

6) Satoshi Hayashi : Current Procedure for Earthquake Resistant Design of Harbour

Structures in Japan, Proceedings of the First Joint Meeting, U.S. -Japan Panel on Wind and Seismic Effects, UJNR, May 1969.

7) Satoshi Hayashi : Comments on Revision of Earthquake Resistant Design Criteria for

Harbour Harbour Structures in Japan, Proceedings of the Second Joint Meeting, U.S.-

Japan Panel on Wind and Seismic Effects, UJNR, May 1970.

8) Ministerial Ordinance on Engineering Requirements on Port and Harbour Facilities

and Its Application, The Japan Port and Harbour Association, February 1975, 94p.

VIII-49
Classif icat ion Region

Regional

seismic

coef f icienC

First region

Hokkaido (Nemuro, Kushiro, Hidaka)

Kanto (Chiba, Tokyo, Kanagawa)

Chubu (Shizuoka, Aichi)

Kinki

0.15

Second region

Hokkaido (Ishikari, Iburi,

Shiribeshi, Hiyama,

Oshima, Rumoi)

Tohoku

Kanto (Ibaragi)

Chubu (Niigata, Toyama, Ishikawa,

Fukui)

Shikoku

Chugoku (Tot tori, Okayama,

Hiroshima)

,

Kyushu (Oita, Miyazaki)

0.10

Third region

Hokkaido (Soya, Abashiri)

Chugoku (Shimane, Yamaguchi)

Khushu, Okinawa (Fukuoka, Saga,

Nagasaki, Kumamoto, Kagoshima,

Okinawa)

0.05

Table 1 Regional seismic coefficient

VIII-50
Classification 1st kind 2nd kind 3rd kind

Factor 0.8 1.0 1.2

Table 2 Factor for subsoil condition

Thickness of quaternary

deposit Gravel Sand or clay Soft ground

less than 5 m 1st kind 1st kind 2nd kind

5 - 25 m 1st kind 2nd kind 3rd kind

more than 25 m 2nd kind 3rd kind 3rd kind

Table 3 Classification of subsoil condition

Classification

of structure Characteristics of structure Importance

factor

Special

class

'I'n ^ of n1 r» t* ol-LU(_v-LlLC 1 1 1*0 Vilido a c 0c iX gllX on ^ f i r» a n f /^Vi a a r* f 0 t*— i. X^dil 1- ^11 dL dV, l-C L istics described by items of (1) - (3)

in A class

1.5

A class

(1) If the structure is damaged by an earthquake, a large number of human

life and property will possibly be lost.

(2) The structure will perform an important role on the reconstruction,

work of the region after an earthquake

(3) The structure handles a hazardous

or a dangerous object, and it is feared

that the damage on the structure will

cause a great loss of human life or property.

(4) If the structure is damaged,

economical and social activity of the

region will be severely suffered.

(5) If the structure is damaged, it

is supposed that the repair work of

it is considerably difficult.

1.2

Classification

of structure

Characteristics of structure Importance

f artor

B class

The structure is other than Special,

A and C classes. 1.0

C class

The structure is small and easy for

repairment, excepting that in Special

and A class.

0.5

Table 4 Factors depending on importance of structure

VIII-51
Fig. 3. Earthpressure acting on a vertical wall

p = ( w+r -h )K ( 1)

K= cos'( i>±<p-d )

cos

CO tC=+ tan i<t>±8±<f) + sec{ cosf sin( ^-^^ )

where; p : intensity of lateral earthpressure in earthquakes (t/m2)

w : intensity of uniform load on the ground surface (t/m^)

k : seismic coefficient

h

K

V

s

d

c

angle of internal friction of sandy soil (")

for general case ..,30°

for particularly good backfill ...40°

unit weight of soil (t/m^; buoy unit weight should be

used below water level and the following are the standards:

above water table in backfill ...1.8t/m3

below water table in backfill ...l.Ot/m^

depth from the ground surface (m)

coefficient of lateral earthpressure

angle between wall surface and the vertical (°)

angle of friction between soil and wall (°); usually |5| 15°

angle given by the following equations ;^=tan~l k ovO =tan~lk'

angle between failure surface and horizon (°)

VIII-52
Fig. 4. Coefficient of active earthpressure

by Mononobe-Okabe's formula

Kind of stress

SS 41, SM 41

SMA 41

SS 50 SM 50

Axial tensile

stress

(for net section)

1,400 1,700 1,900

Axial compressive

stress

(for gross

section)

£: effective

buckling length

(cm)

r: radius of

gyration for

gross section

(cm)

(a) ^<20, 1,400

(b) 20<^<93

1,400 - 8.4(^-20)

(c) 93^f

12,000.000

(a) ^^17 1,700

(b) 17<y<86

1,700-11.3 (y -17)

(c) 86^^

r

12,000,000

(a) y^l5 1,900

(b) 15<^<86

1,900-13( ~~ 15)

(c) 80^^

12,000,000

6,700+ (^/r)^ 5,700 + (^e/ rf 5,000+ {e/rf

Bending stress

(1) tensile stress

for net section

(2) compressive

stress for

gross section

1,400 1,700 1,900

Shearing stress

(for gross

section)

800 1,000 1,100

(unit : kg/cm2)

Table 5 Allowable stresses of structural steel

VIII-53
Kind of stress

SS 41, SM 41

SMA 41, STK 41

Axial tensile stress

(for net section) 1,400 1,900

Axial compressive stress

(for gross section)

bame as snown in laoj-e j

Combination of axial

compressive force and

bending moment

<7 a. +

be ^ 1.0 a a, —

ca ba

Shearing stress

(1) with stiffener

(2) without stiffener

R/t<125

800-0. 019(R/t)2

200>R/t>125

75,000/ (R/t)-90

500

R/t<95

1, 100-0. 044(R/t)2

200>R/t>95

75,000/(R/t)-90

600

(unit : kg/cm'^)

Notation o

f.'. compressive stress of axial compressive force

"be : compressive stress of bending moment

<'ca • allowable axial compressive stress

"ba '• allowable bending compressive stress shown in Table-5

R : outer radius of steel pile (cm)

t : thickness of steel pile (cm)

Table 6 Allowable stresses of steel pile materials

Kind of steel material Allowable stress

SY 24

SY 30

SY 40

1,400

1,800

2,400

(unit : kg/cm2)

Table 7 Allowable stresses of steel sheetpile materials

Kind of reinforcing bars SR 24 SR 30 SD 24 SD 30 SD 35 SD 40

Tensile stress 1,400 1,600 1,400 1,800 2,000 2,100

Tensile stress for

fatigue loading 1,400 1,600 1,400 1,600 1,800 1,800

(unit : kg/cm2)

Nation 1) Diameter of bars is less than 32 mm

2) When design standard strength "ck is less than 180 kg/cm2

deformed bars (7sa<l,600 kg/cm2

round bars ffsa<l»200 kg/cm2

Table 8 Allowable tensile stress of reinforcing bars og

VIII-54
/

1

1-

J

-V t \

" 1

»R-

/

If)

o/

1

' m/ u

<3 In

% ! a

11 II Hi —rC

'Z. o

to o

cvi

Kind of

stress

Flexural

compressive

stress

Shearing

stress

Bond stress

Flexural

tensile

stress

Items

beam

slab

*^a2 only shear force

round bars

deformed bars

reinforced concrete

'ck

180 240 I 300

'ca

6

8

17

7 14

20

8

16

8

10

22

9

18

Plain concrete

4

'ca

7

-'ck

"ca M

design standard compresive strength

design standard tensile strength

allowable stress

allowable shearing stress without calculation of

diagonal tension bars

allowable shearing stress with calculation of diagonal

tension bars

<uu

3

«3

CO

01ua

-cu

u

to

(u cd

>

•H

cn

cn

•j-i o

c

01 •HO

•H

M-l

<4-{ o;ou

0)

o

B

o

so

60

•H

U4

Table 9 Allowable stresses of concrete

VI 1 1- 5

5
JSCE SPECIFICATIONS FOR EARTHQUAKE RESISTANT DESIGN

OF SUBMERGED TUNNELS (197 5)

by

Eiichi Kuribayashi

Chief, Earthquake Engineering Research Section

Public Works Research Institute

Ministry of Construction

and

Hajime Tsuchida

Chief, Earthquake Resistant Structure Laboratory

Port and Harbor Research Institute

Ministry of Transport

ABSTRACT

In response to the request from Ministry of Construction and Ministry of Transport,

the Japan Society of Civil Engineers has concluded in March 1975 the final draft of the

Specifications for Earthquake Resistant Design of Siibmerged Tunnels. The writers of this

paper have worked on the drafting of these Specifications in cooperation with colleagues

of the Public Works Research Institute and Port and Harbor Research Institute during the

last four years.

The draft of the Specifications was adopted as Specifications for Earthquake Resistant

Design of the Proposed Tunnel across Tolyo Bay. This paper presents the principal provisions and articles of the draft of the Specifications, and contains the following five

chapters; General, Investigation, Earthquake Resistant Design, Dynamic Analyses, and

Preservation and Countermeasure in Earthquakes

.

Key Words: Aseismic Design Criteria; Design Provisions; Earthquakes; Specifications;

Structural Engineering; Tunnels.

VIII-56
Foreword

The Committee of Earthquake Resistant Design criteria for Siabmerged Tunnels , chaired

by Professor Shunzo Okamoto, was established by the Japan Society of Civil Engineers under

a contract with the Public Works Research Institute, Ministry of Construction in July, 1971.

The contract had been supported by the Institute in addition to The Second Construction

Bureau for Port and Harbor, Ministry of Transport since 1972.

Since the initiation of this work, the committee has met nine times with the Sub

Committee, who drafted the specifications and have met forty-four times. Through 1971 to

1972, the dynamic behavior of siabmerged tunnels was examined, and then the provisions of

the Specifications were drafted on the bases of these results.

Prior to developing the specifications, the application of earthquake engineering and

design practices to submerged tunnels both in Japan and abroad, and the general observation

of submerged tunnels during earthquakes were comprehensively studied. The Committee, however, adopted the final resolution as a draft to the Specifications with due consideration

that lack of sufficient examples of design applications existed. This draft will attempt

to establish a meaningful specification for the future and thus provide needed experience

on their application of earthquake resistant design.

The safety and stability of the submerged tunnels are important and therefore, the

skeleton of the specification draft should be applied toward the design and construction of

tunnels with care. The draft will, however, be an important guide in the construction and

the design of tunnels.

Finally, the members of the Committee which participated in the draft of the Specifications included the Ministry of Construction, the Ministry of Transport, and the clerk

of the Committee of Japan Society of Civil Engineers. Their cooperation is deeply

appreciated.

VIII-57
Chapter 1 General

1.1 Scope

The provisions in the Specification apply to earthquake resistant design of submerged

tunnels

.

1.2 Conformation to Specifications

In regard to matters which are not specified herein, the following Specifications

shall apply in accordance with the types of structures and facilities considered.

Standard Specifications of Concrete,

Japan Society of Civil Engineers

Specifications for Design and Construction of Prestressed Concrete,

Japan Society of Civil Engineers

Specifications for Earthquake Resistant Design of Civil Engineering Structures,

Japan Society of Civil Engineers

Specifications for Earthquake Resistant Design of Highway Bridges (January, 1971)

,

Japan Road Association

Design Standards on Seismic Effort for Port and Harbor Structures,

Port and Harbor Association of Japan

Uniform Building Code,

1.3 Definitions of Terms

The following definitions shall apply only to the provisions of these Specifications.

Submerged Tunnel; A total structural system which are composed of submerged structures,

approaches and ventilation towers

.

Submerged structure; A portion of the tunnel which is under water on a water table. Those

portions are constructed in the following sequence.

1st: The elements are fabricated in yards or docks

2nd: The elements are floated and towed to the construction site.

3rd: The elements are placed on a trench prepared along the site.

4th: The elements are connected together under the water tables.

5th: Over-all elements are covered with rock and soil.

Approach; Under-ground structures or exposed structures which lead from the submerged

structure to airial portions.

Ventilation tower; A structure which is located at the intermediate portion of the tunnel

for the purpose of construction and ventilation of the tunnel.

Seismic Deformation Method; A method for earthquake resistant design of the tunnel, in

which the ground displacements at the level of an axis of the tunnel are assumed

to apply seismic effects to the siibmerged structure.

(Fourteen additional terms, other than above, are herein omitted).

VIII-58
Chapter 2 Investigation

2.1 General

For earthquake resistant design of a submerged tunnel, detail investigations shall be

made on earthquakes and earthquake ground motions; geology and soils; materials, and types

and details of structure; preservation and counter-measures in earthquakes; and a post- earthquake inspection program.

2.2 Investigation of Earthquakes and Earthquake Ground Motions

2.2.1 Earthquakes

Investigation on earthquakes are presently being conducted in order to collect data

on seismic activity in an area of a proposed submerged tunnel site. The data will be

utilized for earthquake resistant design and for preservation and counter-measures in

earthquakes

.

2.2.2 Earthquake Ground Motions

Field observations on earthquake ground motions are being conducted in order to obtain data necessary to estimate earthquake response of soil strata and structures.

Such estimations shall be taken into consideration for earthquake resistant designs

and on preservation and counter-measures in earthquakes

.

2.3 Investigation of Geology and Soils

2.3.1 General

Investigation on geology and soils in relationship to earthquake resistant design

is divided into two categories; (1) preliminary investigation and (2) site investigation. The preliminary investigation is required in order to collect infomation on geology and soils in the tunnel locality, preceding the site investigation for making

the site investigation efficient and effective. The site investigation is conducted

in order to obtain all the necessary data for design and construction of a siibmerged

tunnel

.

2.3.2 Preliminary Investigation

In the preliminary investigation the following information shall be collected:

1) Topographic map; also submarine topographic map when it is necessary

2) Geological map

3) Soil profile

4) Soil map

5) Boring log

2.3.3 Site Investigation

In conducting the site investigation, the following items should be determined;

possible structural types of a submerged tunnel, soil condition, and other related

factors. In considering these items the following viewpoints shall be made of the

site investigation.

VIII-59
1) Data necessary for determination of design earthquakes

2) Data necessary for structural design

3) Data necessary for examination of stability of soils during earthquakes

Items that are included in site investigations depend on the proposed tunne; however,

the following items shall be included in the site investigation.

1) Boring, sounding (normally standard penetration tests) , and sampling

2) Laboratory tests of samples

3) Measurement of seismic wave velocities with boreholes.

4) Measurement of density of soils

5) Microtremor observation

2.3.4 Test Procedures

Field and laboratory tests shall be conducted in accordance with procedures specified in the Japan Industrial Standards or the Standard Specifications of the Japanese

Society of Soil Mechanics and Foundation Engineering.

2.3.5 Configuration and Depth of Base-rock

In the site investigation configuration and depth of base-rock surface shall be

surveyed. The base-rock surface means an interface between base-rock and surface

layer.

2.4 Investigation of the Engineering Property of the Soil and Surface Layer

2.4.1 Engineering Property of the Soil and of the Surface Layer

The following are the important properties to be considered for earthquake resistant

design.

1) Density

2) Elastic moduli; Yoving's modulus and shear modulus

3) Poisson's ratio

4) Strength parameters; angle of internal friction and cohesion

5) Coefficient of subgrade reaction

6) Strength of soils under dynamic loading

7) Velocities of seismic waves; velocity of longitudinal waves and velocity of

transverse waves

8) Dynamic characteristics of subground

These properties shall be determined directly by observation or tests. Only when it

is very difficult to be obtained directly, the properties can be determined indirectly

using the results from standard penetration tests or other observation tests.

2.4.2 Strength of Soils under Dynamic Loading

Dynamic strength-test of soils, to examine the soil behavior under dynamic loading,

is desirable.

2.4.3 Measurement of Velocities of Seismic Waves

Velocities of seismic waves shall be measured at the proposed tvinnel site.

VIII-60
2.4.4 Measurement of Microtremor

For estimation of the dynamic characteristics of the surface layer, at the tunnel

site, microtremors shall be measured and analysed.

2.4.5 Damping Factor of the Surface Layer

Damping factors of surface layer, which are applied in the earthquake response

calculations and dynamic model tests shall be carefully determined based on results

of field and laboratory tests.

2.4.6 Properties and Strain of the Soil

Properties of the soil and surface layer, such as elastic moduli, velocities of

seismic waves, and damping factor, depend of the strain in the soil. The difference

between the strain in the soil under test or observation and the strain in the soil

expected during earthquakes shall be considered when the properties are used in

earthquake resistant design.

2.5 Ground Failure

A submerged tunnel is supported by the surrounding soil which is usually soft, and

thus the stability of a submerged tunnel depends largely on the behavior of the ground

during earthquakes. Engineers concerned with the design and construction of a submerged

tunnel shall have appropriate knowledge of the ground failure due to earthquakes.

2.6 Seismic Stability of the Soils

2.6.1 Stability of Sandy Soil

Liquefaction of soils around a submerged tunnel shall be avoided. When the soil

is estimated to liquefy during earthquakes the soil shall be improved. Liquefaction

potential shall be examined based on liquefaction case records from past earthquakes

and results of research on liquefaction.

2.6.2 Stability of a Cohesive Soil

When soil at a tunnel site is cohesive, it shall be checked such that a change in

the soil strength and deformation due to earthquake will never affect the stability

of a submerged tunnel.

2.7 Investigation of the Materials and Structural Types

Relative to materials, types and details of a submerged tunnel, tests and investigations shall be conducted on the following;

1) Structural concrete

2) Structural steel

3) Watertightness of elements

4) Types and materials of the joints

5) Types of foundations

6) Preparation of trench bottom and backfilling

VIII-61
2 . 8 Investigation of the Preservation and Proper Countermeasures in Earthquakes

For the preservation and countermeasures of tunnels in earthquakes, traffic control

systems of the tunnel shall be examined, and the interaction of the function among these

systems shall be sufficiently considered.

Chapter 3 Earthquake Resistant Design

3.1 General Principle

Every partial structural system shall be designed by the seismic deformation method

and the seismic coefficient in accordance with the provisions in Chapter 3 "Earthquake

Resistant Design". Also the total structural system shall be designed by using the results

of the dynamic response analysis with regard to the influence of the surrounding topography

and geology in accordance with the provisions in Chapter 4 "Dynamic Analysis".

Equipment shall be provided for earthquake control and examination, and shall be used

in accordance with the provisions in the Chapter 5 "Preservation and Countermeasures for

Earthquakes".

3.2 Design Requirements

1) Not only the submerged structure but also the total structural system, including

the effects of the surrounding topography and geology, shall be designed to provide

sufficient stability against seismic disturbances.

2) Every structural system of the submerged tunnel shall be designed in accordance

with the displacements of the surrounding ground during earthquakes and design

seismic coefficients. The structural system shall be designed and the results

examined by the dynamic response analyses.

3) The structural system in which the rigidity changes, i.e. joints, hinges and other

parts, shall be designed with regard to the effects the change in rigidity has on

the system and seismic resistance of the submerged tunnel

4) For the preservation and countermeasures of the tunnel during earthquakes, the

reliability of the operation of the systems of the submerged structure, the approach,

the ventilation tower and other equipment shall be considered.

3.3 Loadings and other Conditions in the Earthquake Resistant Design

3.3.1 General

1) The following Loadings and Conditions shall be taken into account in earthquake

resistant design. The Appropriate Loadings shall be selected from this list on

the basis of the location and the type of the structure.

a) Dead Loads

b) Earth Pressures

c) Hydrostatic Pressures

d) Buoyancy or Uplift

e) Live Loads

f) Effects of Consolidation and Settlement of the Sub-ground

g) Effects of Temperature Change

VIII-62
h) Effects of Shrinkage due to Humidity in Concrete Structures

i) Other Loadings (Tidal Waves, etc.)

2) The following seismic effects shall be taken into account in earthquake resistant

design.

a) Displacement of the sub-ground or structures in earthquakes

b) Inertia forces due to the dead weight of the structure

c) Earth pressures in earthquakes

d) Hydrodynamic pressures in earthquakes

3) Combination of Loads

Design conditions shall be determined considering the loading conditions a)

through h) , given in the preceding article (1) and in addition the loading

effects h) and i) in accordance with the site conditions.

3.3.2 Ground Displacement in Earthquakes

1) The displacements of the subsurface ground in which the submerged tunnel is

embedded shall be determined in accordance with the provisions given in Section

3.4.2.1 "Ground Displacement in Earthquake Resistant Design". The ground

Displacement for the design of siibmerged structures shall be taken as the

ground displacement at the level of the longitudinal axis of the structure.

2) The plane where the ground displacements are applied, shall be taken as the

horizontal and the vertical plane on the axis respectively.

3.3.3 Inertia Forces (omitted)

3.3.4 Earth Pressures due to Earthquakes (omitted)

3.3.5 Hydrodynamic Pressures due to Earthquakes (omittedX

3.3.6 Soil Layers Where the Bearing Capacities are Neglected in

Earthquake Resistant Design (omitted)

3.4 Design Earthquake Ground Motion

3.4.1 General

1) The ground Displacement and the design seismic coefficient shall be taken into

account in the design of the submerged tunnel

2) The submerged structure shall be designed by the seismic deformation method and

the seismic coefficient method.

3) Ventilation towers and other structures can be designed by the seismic coefficient method.

3.4.2 Design Earthquake Ground Motion

3.4.2.1 Ground Displacement in the Earthquake Resistant Design

The ground displacement, in the earthquake resistant design, shall be evaluated

on the basis of the ground displacement considering the nature of the earthquake

ground motions and the soil conditions.

VIII-63
3.4.2.2 Base Rock Accelerations in Earthquake Resistant Design

1) The horizontal base rock acceleration, in earthquake resistant design, shall

be evaluated in accordance with the intensity of the earthquakes at the

construction site and the importance of the submerged tunnels.

2) The vertical base rock acceleration, in earthquake resistant design, shall be

taken as one half of the horizontal base rock accelerations.

3.4.2.3 Design Seismic Coefficient in the Seismic Coefficient Method (omitted)

3.5 Earthquake Resistant Design of Submerged Structures

3.5.1 General

The submerged structure shall be principally designed by the seismic deformation

method. However, the seismic coefficient method can be applied to the design of the

transverse tunnel sections and the examination of possibility of the sliding of the

submerged tunnel.

3.5.2 Seismic Deformation Method

The submerged structure shall be designed in accordance with the Seismic Deformation

Method.

3.5.3 Seismic Coefficient Method (omitted)

3.6 Earthquake Resistant Design of Ventilation Tower

3.6.1 General

The earthquake resistant design of the ventilation tower shall provide appropriate

stability against seismic disturbances for the total structural system of the submerged

tunnel, considering the conditions for the connection with the submerged structure.

3.6.2 Earthquake Resistant Design of the Ventilation Tower

The ventilation tower can be designed by the seismic coefficient method. The design

seismic coefficient, in the seismic coefficient method, shall be determined in accordance with the provisions in Section.

3.4.2.3 "Design Seismic Coefficient in the Seismic Coefficient Method"

The directions of the inertia forces and the design method shall be determined

in accordance with the provisions in individual related Specifications, according

to the type of the ventilation tower considered and the conditions of the construction site.

3.7 Earthquake Resistant Design of Approaches

3.7.1 General (omitted)

3.7.2 Earthquake Resistant Design of Approach (omitted)

3.8 Stability of Subgrounds

3.8.1 General

The stability of the ground around the submerged structure shall be evaluated.

VIII-64
3.8.2 Stability Analysis of Ground

The Stability of the ground can be analyzed by the seismic coefficient method

assuming a slip plane. Also the submerged structure shall be analyzed to avoid the

critical damages due to the ground displacement evaluated by the dynamic analyses.

3.8.3 Stability of the Soil-use for Submerged Structure Fill

1) Liquefaction

A soil which is prone to liquefaction during earthquakes shall not be used for

soil fill for submerged structures. The estimate of liquefation shall be

evaluated in accordance with the provisions given in .section 2.6.1 "Stabiltiy of

Sandy Soils".

2) Examination of Sliding

The stability of the sliding of submerged structures in the transverse direction

shall be analyzed in accordance with the provisions in Section 3.5 "Earthquake

Resistant Design of Submerged Structure".

3.9 Allowable Stresses

3.9.1 General

The allowable stresses of the materials, and the increase of the allowable stresses

for earthquake resistant design shall be determined in accordance with the structural

types and the importance of the structures.

The stresses shall be required to be based on the various Specifications

and the consideration of the assumptions in the design and the procedure and operation

in the construction.

3.9.2 Allowable Stresses of Concrete

1) The allowable stresses of concrete, for Use in submerged tunnels, shall conform

to the provisions specified in the following:

Standard Specifications of Concrete,

Japan Society of Civil Engineers

2) The allowable stress of concrete placed in water shall be determined on the

basis of experiments of simulated field conditions, because of the lack of

reliability of the uniform quality of the concrete.

3.9.3 Allowable Stress of Steel

1) The allowable stress of steel used for sxibmerged tunnels shall conform to the

provisions specified in the following Specifications.

For reinforcing Bars;

Standard Specifications of Concrete

,

Japan Society of Civil Engineers

For Structural Steel, Except for Reinforcing Bars;

Standard Specifications for Highway Bridges, Part of Steel Bridges,

Japan Road Association

VIII-65
2) In the case of minor cracking of the concrete, the allowable tensile stresses

shall be designated in accordance with the effect of cracking of concrete.

3.9.4 Increase in the Allowable Stresses in the Earthquake Resistant Design

1) Use of the allowable stresses specified in the provis ions of the Section 3,9.2

"Allowable Stresses of Concrete" and the Section 3.9.3 "Allowable Stresses of

Steel", the increase of the allowable stresses shall be allowed in accordance

with the following combination of loadings;

Table 3.3

Combination of Loading Limitation of Increase of

Allowable Stress

Ordinary Load + Earthquake

Loading

Ordinary Load + Effect of Temp.

Change + Shrinkage + Earthquake 65%

Loading

Note) In this table, the fundamental allowable tensile stress shall

confojrm to the provisions of Section 3.9.2 "Allowable Stresses of

Steel".

2) If the provisions of the Section 3.9.1 "Allowable Stresses of Concrete" and

the Section 3.9.3 "Allowable Stresses of Steel" are not used in the evaluation

of the gross allowable strength of the section of the structural member, the

allowable stresses for ordinary conditions and seismic conditions or the safety

factor shall be determined on the basis of the experiments or equivalent measures.

3.10 Earthquake Resistant Design in Detail

3.10.1 General (omitted)

3.10.2 Attached structure (omitted)

3.10.3 Pile Foundation (omitted)

3.1014 Ventilation tower (omitted)

Chapter 4 Dynamic Analysis

4.1 General

The dynamic analysis for the design of a submerged tunnel is divided into the dynamic

analysis for the partial structural system and the dynamic analysis for the total structural

system.

4.2 Earthquake Response Analysis

4.2.1 General

The earthquake response analysis for the submerged tunnel shall be conducted for

the individual items of the structural system, such as the surrounding ground, the

VIII-66
submerged structure and the ventilation tower, in addition to the total structural

system. The analysis shall consider the dynamics of the soil-structure interaction and

the characteristics of earthquake ground motions by using the method which can simulate

the dynamic response of the submerged tunnel as close as possible.

4.2.2 Method of Earthquake Response Analysis

The earthquake response analysis can be conducted by using one of the following

two methods

.

(A) method using the averaged response spectrum,

(B) method using the original record of earthquake motions.

4.2.3 Mechanical Model

The mechanical model, for the earthquake response analysis, shall be individually

designated to represent such structural systems as the ground, the submerged structure,

the ventilation tower and the submerged tunnel of the total system and taking into

account if necessary the effects of the existence of the water. The dynamic properties

of the ground, such as natural period, vibration mode, damping characteristics and the

dynamic characteristics of the submerged structure and the attached structures, and

the dynamic interaction among the ground and the structures shall be taken into

account in representing mechanical mode.

4.2.4 Input Earthquake Motion

The input earthquake ground motion shall be designated by the maximum acceleration

and the characteristics specified in the provisions of Chapter 2 "Investigation".

The input earthquake ground motion shall be applied to the base rock.

4.3 Dynamic Model Test (omitted)

4.4 Designation of Safety (omitted)

Chapter 5 Preservation and Countermeasures in Earthquakes

5.1 General

For the purpose of preservation and countermeasures in earthquakes , earthquake

equipment shall be installed in the submerged tunnel. The security and the stability of

the tunnel are guaranteed completely by using the equipment and also conducting inspections.

5.2 Preserving Equipment

For the purpose of the preservation and countermeasures in earthquakes , the following

equipment shall be installed. This equipment shall be guaranteed to have normal operation

during earthquakes and have periodical inspections

.

1) Sensor, recorder and system of notification for earthquake ground motion

2) Sensor, recorder and system of notification for the level of tidal wave and tsunami

3) Traffic control system during and after earthquake

4) Sensor, recorder and system of notification for settlement and spill water

5) System for evacuation and induction

5) Emergency power plant

VIII-67
7) System for inspection of traffic

8) Drainage and system of protection for inundation

9) Other (strain-meter, stress-meter, etc)

5.3 Preserving Operation

By employing the preserving equipments, the countermeasure for the security of traffic

and the safety of the structure shall be applied in accordance with the intensity of earthquakes at the location of the tunnel.

5.4 Inspection for preservation

The tunnel structure shall be maintained by inspection, for proper preservation. The

inspection shall be periodically conducted, and also the temporary inspection shall be

conducted during earthquakes of the strong intensity.

5.5 Investigation after earthquakes

The administrator of the submerged tunnel shall immediately conduct an investigation

of every portion of the tunnel after earthquakes of high intensity. The results of the

investigation and the proper countermeasures shall be employed and recorded.

VIII-68
JOINT RESEARCH PROGRAM UTILIZING THE

LARGE SCALE TESTING FACILITIES (FREE DISCUSSION)

by

Makota Watabe

Senior Staff Scientist

Structures Division

Building Research Institute

1 and

Masaya Hirosawa

Senior Staff Scientist

Structures Division

Building Research Institute

and

Shinsuke Nakata

Researcher

Structures Division

Building Research Institute

ABSTRACT

Recently the United States and Japan have made progress in aseismic design of R.C.

Structures based on the testing of R.C. members, in which these tests have shown the importance of ductility in columns. However there are a few seismic studies based on the

loading tests of full size structures. Fortunately, in Japan there is the greatest loading

test facilities in the world, with new larger facilities now under construction. By using

these facilities full size asemismic tests can be conducted. This proposal provides a plan for the testing of reinforced concrete structures with shear walls, which should be

the first step toward developing an international aseismic code.

Key Words: Dynamic Testing; Laterlal Load Simulation; Models; Shake Tables; Testing.
Recent Research Progress

R.C. structures are generally recognized to have sufficient earthquake resistance.

However in 1968, the Tokachi-Oki earthquake induced extensive damage to school buildings

in Japan. In 1971, the San Fernando earthquake induced severe damage to hospital buildings

in the USA. Examination of the earthquake damage indicated that many R.C. columns sus- tained severe damage, particularly shear failures. Thus, the problem of the need for ductility

of R.C. column was reconfirmed. Therefore, since 1973, a research project was organized by

the Ministry of Construction which included tests of about 200 R.C. column specimens. The

following lists some of the results obtained from these tests;

(1) All types of failure modes of the coliamns were defined precisely from the tests.

They consist of shear diagonal tension failure, shear compression failure, shear tension failure, bond splitting failure, and buckling of the main bars. Empirical equations relative to strength associated with each mode were then established.

(2) Load-deflection curves of all type of columns were standardized and the limitation

of web reinforcement (quantity, shape and arrangement) for ductile members was determined.

Further Problem

At present many experimental studies have been conducted on columns , beams and shear

walls. However, there are many fields still unexplored. For example, the seismic tests

(static and dynamic) of full size structures with slabs, shear walls and foundations need

investigations. These test results would provide insight on the difference between the

behavior of a structure and unit member, especially the anchorage of the main bars or ground effects on the walls, etc. Previous studies on the structural damage due to

earthquake were based only on member tests in the laboratory, and the behavior of entire

structures had to be based on many engineering assumptions, which may lead to some errors.

Full size structural testing should be made in order to avoid such errors.

Planning of Test

a) Testing Facilities

In Japan, there are large dynamic and static testing facilities and many others

now under construction. Two such facilities are described below;

(Large Scale Earthquake Simulator)

This simulator system is housed in the National Research Center for Disaster

Prevention and located in Tsukuba Research and Education City (that is now under

construction) . The performance of this simulator is shown in Table 1. The test

platform is 15 by 15 meters in plan, which allows free motion in two directions.

The effective weight on the platform is 500 ton (horizontal motion) , and 200 ton

(vertical motion) . The test platform is actuated by a 360 ton capacity hydraulic

actuator at a maximum acceleration of 500 gals and a maximimi velocity of 37 cm/sec

and a maximum displacment of ± 30 mm under dynamic conditions. Fig. 1 shows the

relationship between acceleration, frequency and weight.

IX2
(Large Size Loading Facility)

This facility is for both static and dynamic full size structural tests using

hydraulic serve actuators, which are not under construction. The plan and section

of this facility are shown in Fig. 2. The facility consists of two blocks A and B,

and a wall for horizontal reaction which is located along the periphery of the blocks.

B block has two reaction walls in two directions. All the testing systems, loading,

vibration and measurement are directly controlled by a computer systen in this facility

Fig. 3 shows an example of the testing capacity of this system,

b) Description of Test Structure

As a first stage of the testing project, it is proposed that a four story building

with shear walls be tested, as shown in Fig. 4. The type of shear walls will include

normal types, precast walls and slip walls. Conducting of this test will permit

solution of such problems as the actual behavior of shear walls, effects of the founda

tion, soil conditions and the mechanism at structural yielding. The test structure

as shown in Fig. 4, is one example, other types of buildings however, should be included in phe project, i.e. structures without shear walls and with short columns or

long columns, long span structures and many types of ground conditions.

Conditions

Important buildings and materials must be protected during strong earthquakes. From

this viewpoint, therefore, our first study should be the development of full size aseismic

tests using Japanese large scale facilities. Systematic full size tests in cooperation

between the U.S.A. and Japan will permit a better understanding of factors in order to

establish an international standard aseismic design code.

IX3
Table 1 Capacity of Vibration Table

Tabic ISm X 15m , weight 160 tons

Vibration system Electric oil compression

Max. power 360 tons tons X 4)

Max. weight ^00 tons (horizontal)

200 tons (vertical

)

Ma N • jnp 1 i t ude :iO mm

Max. velocity 37 cn/sec

Max. acceralation 0.55 g (horizontal

)

1.0 g (vertical)

Frequenc y 0. 1-50 HZ

Structure 24m X 42.5m H=12.2m

Crane 30 tons and 5 tons

Frequency (Hz)

FIG 1 VIBRATIONAL CAPACITY

H 1^... .P.... H.

" U 3.3 'IJ

:3J

St

" .'. " .' . '^ TT

TffTTirriTr'TTT " '

' 1 1 ! f ''

' XT

I r--T I \U\

\

'

I L..J

I r"T

I '

I

I r'"i

' L j

I

I L.J

I r---

J I—

J

n

S 0 Li a:

TT

ni u l.tl ^ LI n Lyfi^l^'^li,*

REACTION WALL

1 tJT»

SECTION

051*4 .2J —

r

7.S5

1 i T

i _

;

3

-ra--------

.J

i: ::::j:l::::t:i !:!:H^:!::::t:!

-WVv.

: VrA^. v;-^ : S^aX

--i:[T^TT:j-:--^:f-:-;-rs:<r

1. O > i - - i- r • i i-B- ; j

• J

L-.J r""i

axil

oaJI m I n I ii I to I 13S I|aj5

I M I 1^ I M ^ |ll 111 1* III III lf.\\\

l3ttg

PLAN

FIG 2 LARGE SCALE ASEISMIC FACILITY

IX-4
IX5
EARTHQUAKE DISASTER MITIGATION:

A Joint Research Approach

by

Charles C. Thiel

and

John B. Scalzi

Program Managers, Earthquake Engineering

Research Applications Directorate (RANN)

National Science Foundation

ABSTRACT

This paper discusses the current earthquake research being undertaken in the USA

under sponsorship of the NSF. In addition, possible cooperative research studies between

the USA and Japan are described.

Key Words: Buildings; Earthquakes; Research; Structures; Structural Engineering.

IX-6
Introduction

Numerous recent publications attest to the seriousness of the hazard presented by

earthquakes to the peoples and economies of the United States and Japan. The damaging

quakes of this decade are grim reminders of the severe impacts of the 1923 Tokyo and 1906

San Francisco events. In the intervening years these cities have grown, concentrating substantially greater populations likely to experience earthquakes of comparable intensity.

Urbanization combined with industrialization are increasing our vulnerability not only to

calamitous life losses but to major economic disruption and dislocation through lost industrial capacity.

These are not problems without potential resolution. It is a tenet of engineering that we can devise strategies at some cost that will reduce the potential for disruption. It is

this belief that drives the research process to discover and develop better economic design

and construction practices. Our joint national interests in this area are consistent and

compatible: to control the consequences of earthquake occurrences through limiting life

loss and injury, property damage and social disruption.

The United States program of research in earthquake mitigation is directed at the

development of

(A) economically feasible design and construction methods for building earthquake

resistant structures of all types, and for the identification and repair of

existing hazardous structures;

(B) an operational program for predicting damaging earthquakes and their physical

effects in the seismically active regions of the United States;

(C) procedures for integrating data and information about seismic risk with ongoing land use planning and regulation activities;

(D) procedures for identifying, evaluating and accurately characterizing seismic

hazards in earthquake-prone regions;

(E) improved understanding of the social and economic consequences of individual

and community decisions on earthquake-related issues, emphasizing risk control, pre-event planning, issuance of warnings, provision of emergency services, rescue, recovery and redevelopment; and

(F) methods and procedures for the control or alteration of seismic phenomena

and their effects on existing structures.

The achievement of these objectives will require the concerted effort of many professionals

working in concord.

The problem posed by earthquakes to the U.S. are comparable in most ways to those

posed to Japan. Both our countries have embarked on extensive programs of research in

engineering, seismology, geology and the social sciences to develop earthquake mitigation

procedures consistent with the above objectives. It is certainly true, in the U.S., and

almost surely true in Japan that the magnitude of the research problems we face are significantly greater than the man power and financial resources we can bring to bear to resolve

them. Under the auspices of the U.S. - Japan Panel on Wind and Seismic Effects we are now

IX-7
meeting for the seventh time to discuss our mutual problems and accomplishments. This is a vital element in keeping ourselves jointly informed of the work being pursued in our res- pective countries. We believe that it is now time to escalate our activities to the next

logical plateau: the conduct of joint and complementary research projects.

First an observation. This Panel has by the very nature of its membership been

strongly oriented toward engineering problems. But the problems we face in earthquake

disaster mitigation are not simple engineering, or for that matter seismological , economic

or political. There appear to be eight approaches to limiting earthquake impact, four each

of a physical and social type:

Control the event by prevention or modification of the event;

Anticipate the event so that remedial actions may be taken;

Identify the seismic potential of areas;

Construct facilities so as to perform acceptably during and after the event;

Plan for the warning, response, and recovery to the event;

Distribute the economic risk;

Generate and select alternative physical development plans; and

Adopt and enforce zoning, construction and management standards.

Clearly if our panel is going to focus on the problem of reducing earthquake impacts,

we must extend its membership (or foster companion panels) to take advantage of the experience and capabilities of the other diciplines. The area of earthquake prediction is a vital

area, advancing quickly in our respective countries. This panel could act as a focus for

the exchange of mutually beneficial data and accomplishments. We propose, at the least, to

extend the Panel's activities to include prediction as a sub-panel, possibly meeting separately. In the main, we propose to extend the Panel to address the full range of earthquake

disaster mitigation methods.

As noted above we also feel that it is time for us to begin the pooling of our joint

national research resources to resolve problems of piiblic safety. As noted in the appended

list, the NSF program has made substantial investments in research in a variety of areas.

In keeping with our goal to extend the scope of the Panel's activities, we propose that six

specific areas be targeted for potential joint research:

1. Public policy implications of earthquake prediction;

2. Large scale destructive testing;

3. Design of industrial facilities;

4. Instrumentation at foreign sites;

5. Land use planning;

6. Structural upgrading, repair and retrofitting;

These areas will be discussed individually below:

1. Public Policy Implications of Earthquake Prediction

The issuance of predictions of damaging earthquakes in selected geographic areas may

well be upon us. Forecasts could be one mechanism for reducing potential loss and disruption from earthquakes. With extended lead times, actions could be taken to inspect and

strengthen buildings, upgrade the seismic resistance component of building code requirements

IX-8
so all new structures would be less susceptible to damage, improve land use zoning regulations to limit or prohibit construction in especially hazardous areas and, of course, as the

forecasted event day approaches, plans for partial or complete evacuation could be carried

out.

On the other hand, the negative consequences could be enormous if an extended period

of uncertainty follows a forecast for a damaging earthquake. Px±ilic and Private investment

agents may drastically reduce their construction and development in the area as well as cur- tail production and commercial activity that could trigger an extended showdown in the

local economy which would be reflected in increased unemployment, reduced private income,

shrinking tax base, and increased demands for pioblic services.

Today there is no base of scientific knowledge to provide information on the benefits

or disbenefits of such actions. Limited studies have been initiated in the United States.

They are, however, confined by the time frame in which the research must be accomplished and

by the need to validate policy recommendations by observing the pviblic ' s response to actual

predictions. A joint program of cooperative research could materially reduce the time frame

in which this research is completed and lead to a better utilization of earthquake prediction for the pxablic benefit. We propose a four item cooperative program to achieve the

following objectives:

1. To develop and interpret a comprehensive set of empirically based findings regarding the probable response of organizations and individual citizens to early

credible earthquake forecasts;

2. To develop and apply an effective means of informing organizations and the public

of the findings and their implications;

3. To test in a rigorous fashion the impact of these findings on organizations; and

4. To prepare recommendations for legislative and administrative actions.

To facilitate such a program we propose that a bilateral meeting between appropriate

officials and researchers be concerned in the near future to plan such a program.

2. Large Scale Destructive Testing of Structures

In the area of structural analysis and design we have relied heavily on our technical

capabilities with theory and computers to develop concepts for design of structures to resist earthquake forces. Many of these concepts have evolved from post- inspections of earthquake damage and shake table results. As beneficial as these concepts are there are many

factors which cannot be evaluated by inspection or small scale tests.

Among the parameters for which better data is required are: the three dimensional

behavior of full size structures and individual components subjected to controlled seismic

type forces, the determination of the structural damping characteristics caused by the

various elements in a building, the connections of various structural components and

equipment, the forces acting on various equipment caused by the interaction of the structure

and the equipment, the attachment of non-structural items and other items which are either

too large or too cumbersome to be handled on a shake table or confidently by analysis.

A plan to obtain the required data could be formulated by a Joint U.S. -Japan program

consisting of: (a) large scale testing of existing structures of various types of materials

IX-9
and construction, (b) pseudodynamic tests on full size specimens or structures constructed

to obtain specific data, (c) shake table verifications where required for the pseudodynamic tests or for specimens which may be considered full size.

At the present time a few projects are underway to investigate the behavior of masonry

construction by pseudo-dynamic test methods. Large joint specimens of reinforced concrete

frame and shear walls are being analyzed and tested by pseudo-dynamic test procedures

.

The results of these tests could be verified by full size structural tests to evaluate

the time behavior in a structural system. A program to extend the tests to full size

structures to determine the parameters which can not be evaluated otherwise would be most

desirable and beneficial to both countries.

3. Industrial Buildings

Up to the present time research on buildings has been concentrated on the high rise

type of residential and commercial types because of the hazards to the lives of the occupants. Very little attention has been given or devoted to other types of facilities which

are equally unimportant to the welfare of the community. These are the industrial buildings

which provide the economic base for the existence of the community. The amount of earthquake resistance in these buildings is dependent upon the design engineer of the building

who may or may not have had experience with earthquake design criteria. In general, however,

it appears that most industrial facilities have not been intentionally designed to resist

seismic forces.

The National Science Foundation is currently funding several projects which are related to the industrial sphere of structures and buildings. Among these are: (a) the

analysis and shake table tests of ground storage tanks with varying heights of liquids,

(b) the seismic resistance of the major structural components of fossil fuel power plants,

such as the boiler, the coal handling conveyor system, the stack, the cooling tower, the

piping systems, and the related structures, (c) a study to determine the seismic vulnerability of the electrical distribution system will result in policy decisions for types of

spare parts, readiness of the operational crew, and similar type operational decisions.

As one views the industrial complex of both countries, it is a simple matter to observe

the many types of facilities which will require intensive study and research to develop

economical design procedures which plant managers will be willing to accept. A joint

U.S. - Japan program to investigate these industrial facilities will reduce the duplication

of effort and shorten the period of implementing the research results.

Areas of further research which could be undertaken in consort or separately are:

a. hydroelectric plants and all their appurtenances

b. other components of fossil fuel power which are not already under study such

as turbines, and other pieces of equipment.

c. refineries

d. chemical plants

e. food processing plants

f. warehouses, etc.

IX-10
A plan to attack these problems with the greatest benefit to each country would be to

assign certain projects to each nation and combine the results. Thereby, each nation would

not duplicate the other and the total results would be obtained in much less time, than

would otherwise be possible by each coxantry doing all of the research alone.

4. Instrumentation at Foreign Sites

At any given site earthquakes are relatively rare events. This is fortunate for the

possible affected populus, but leads to inherent difficulty in obtaining strong ground mo- tion useful for extending our design practice. Both the United States and Japan have ac- tive instrumentation programs designed in part to instrument sites of likely activity to

yield data on both ground design and building response. Both our programs are weak, however,

in that we have not necessarily placed these instrumentation networks in the areas of the

world where the data return is likely to be the highest. To date we have amassed a substantial data collection for low magnitude and intensity events . As intensity and magnitude

increase, the recorded data falls off at least logarithmically; indeed data at damaging intensity levels for earthquakes with Richter magnitude above 5.5 are virtually non-existent.

Yet our design considerations are dominated by this magnitude event. There is a definite

world need for near field strong motion data for quakes of magnitude 7 and over. Since

they are infrequent (particularly in the 8.0 range) we must place the networks where these

magnitude events are most likely to occur. This involves the placement of networks outside

of our respective countries. Since we both benefit from this data, we ought to pool our

resources to design, place and maintain these systems.

To accomplish the goal of the timely collection of near field, large magnitude, strong

motion data we propose the joint development of special arrays to be placed outside our

respective countries through the following steps

:

Select regions of the world with sufficiently high seismic activity which are expeditious to develop in such arrays;

Define the nature of the arrays which may be utilized effectively for this purpose;

Evaluate the instrumentation needed to implement these arrays;

Design specific arrays for several of the active areas selected in the first tasks;

Jointly place and maintain the instrument sites;

Regularly exchange data.

5. Land Use Planning/Site Planning

Averting or lessening the potential effects of many geophysical based disaster can be

achieved by regulating the use to which land is put and the materials and methods employed

in the design and construction of physical facilities. For the most part, seismic land use

considerations in the U.S. have been either through the publication of seismic risk maps for

state sized regions, or through the imposition of relatively vague legislative admonitions.

In California's case, they have enacted a statute that requires the consideration of

earthquake hazards in each communities master plan, although there is little agreement on

what such an incorporation should entail as its technical content. Land use is a technique

that offers the opportunity for control not only of earthquake related damage, (shaking,

faulting, landslides, tsunami, etci but also for other geobased hazards. Within the U.S.,

IX-11
it is unlikely that land use criteria based solely on earthquake hazard can be widely adopted.

When floods, landslide, earthquakes and allied hazards are considered, there is some likelihood for implementing technical land use criteria to reduce the public hazard exposure.

Research in this area is embryonic in the U.S. and we conjecture that it is also in Japan.

We propose that a binational working group be established under tha auspices of the Panel

to investigate the most appropriate objectives for joint activities and develop an integrated

research/information exchange program to implement these objectives.

6. Structural Upgrading - Repair and Retrofitting

Most current structural research and that of the past years has concentrated on the

behavior of materials components, structural and systems in order to develop techniques to

improve new construction. Very little, if any research was devoted to the problem of retrofitting for strengthening purposes, or repair after an event. These were left to the

.judgment of the engineer or building official to decide how the building was to be re- paried or strengthened. Their decision was usually based on past personal experience rather

than research results and, in general, these professionals have perfomed well. However,

because many large cities are in potentially high seismic risk zones, and have many old

buildings which are subject to damage if an earthquake strikes, it is imperative that

consideration be given to the problems of retrofitting and repair.

In recent experimental studies of structural components it has become the practice to

test specimens up to a failure point, ut not to total destruction. By this method repairs

are made using epoxies , or grouts and the specimen is retested to determine the new load

resistance capacity. Comparisons of the effectiveness of the repair techniques are made to

determine the best possible method.

A specific project is underway to evaluate the resistance of structures up to four

stories in height, typical of school buildings, and to determine the most cost-effective

method of retrofitting the building. Various materials, such as steel and reinforced concrete with different framing systems are to be studied. The objective for the project is to

develop a methodology for the evaluation of an existing building and to outline the various

methods which may be used to retrofit the specific building.

No doubt, in both countries there are many buildings which could be retrofitted to a higher seismic resistance and consequently provide a greater degree of safety for the

occupants

.

A joint program outlining the types of buildings each country could study would certainly reduce the time and cost of obtaining this most useful information. A subcommittee of

this panel could meet to decide the distribution of activities between countries which

would lead to a common goal.

Conclusion

The time is ripe for our two countries to extend our cooperation in the earthquake area

from one of information to one of joint endeavor. We propose to accomplish this through

two actions. First to extend the participation in the Panel to all of the disciplines en- gaged in earthquake mitigation research and implementation. Second that the Panel foster a

IX-12
series of focused research studies to be undertaken jointly. Obviously such undertakings

will require both the commitment of time and personnel to develop the program and the

allocation of resources to carry it out. We are prepared to act as the focal point for

development of the U.S. participation in such a planning effort.

IX-13
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IX- 16
HIGH WIND STUDY

IN THE

PHILIPPINES

by

Noel J. Raufaste

Federal Building Program Coordinator

Office of Federal Building Technology

Center for Building Technology

Institute for Applied Technology

National Bureau of Standards

Washington, D.C. 20234

ABSTRACT

A review of the National Bureau of Standards three-year high wind study in the Philippines is presented. Accomplishments during the first two years of the study are discussed.

Principal accomplishments include 1) formation of a Philippine Advisory Committee to coordinate local wind research, 2) selection of three field test sites, 3) construction and

instrumentation of six test buildings with wind recording equipment at the test sites,

4) instrumentation of the University of Philippines wind tunnel, and 5) participation in

two international workshops on high winds in Manila.

Key Words: High winds, Philippines, field studies, instrumentation, wind tunnel test.

X-1
Introduction

The National Bureau of Standards (NBS) is developing improved design criteria for lowrise buildings to better resist the effect of extreme winds. The three-year project is

sponsored by Agency for International Development (AID)

.

This research project originated from recognition of a need for additional research to

supplement the limited amount of existing data concerning the effects of wind on low-rise

buildings, especially in developing countries. Many existing criteria for wind loads

design do not make provision for steady and fluctuating wind pressures along the edges of

roofs and walls where flow separations occur. Yet wind pressures along these regions are

one of the primary contributors to building damages. This research and the resultant

development of suitable design criteria and methodologies will reduce losses of structures

and lives in the countries where the criteria are applied.

The Philippines experience the highest worldwide annual frequency of intense tropical

storms. Within the Philippines, the frequency of tropical storms is greatest in Luzon.

Statistics indicate that between 1948 and 1971 the Philippines were exposed to 482 tropical

storms, or an average of 20 per year. As such, these occurrences make the Philippines a natural laboratory to measure wind loads on buldings.

NBS's extreme wind study includes several components. It is based largely on field

work (collection of wind loading data from seven field test buildings) and from wind

tunnel testing of building scale models. The study also includes a review of climatological data from the weather bureaus of the Philippines and two other developing countries,

Bangladesh and Jamaica. Socio-economic, architectural and structural data, from the Philippines and other developing countries will be included in the implementation aspects of

this study. In addition, knowledge about wind effects on buildings obtained in other

countries such as the United States, Japan, Australia and the United Kingdom will be used

as required.

It should be noted that Dr. Richard Marshall, the principal investigator, is responsible for directing the wind research activities as discussed in this report. These

activities are found on pages 3-10.

Background Information

The first step in performing this research was to identify interested organizations,

agencies, universities and other groups within the Philippines, selected Bay of Bengal

countries and several northern Caribbean islands to assess their degree of interest for

possible in-country project participation. On April 27, 1973 the Philippine Advisory

Committee was formed. Professionals from various building related fields were brought

together. The response from the Philippine Government and the private sector numbered

more than 30 scientists, engineers and researchers representing government entities,

professional organizations and private groups. Also included were the USAID Mission to

the Philippines and the U.S. National Bureau of Standards. The Philippine Advisory Committee serves as the focal point for coordinating project activities centering in the

Philippines.

X-2
The committee is composed of the following organizations

:

University of the Philippines

Philippine Atmospheric Geophysical and Astronomical Services Administration (PAGASA)

National Housing Corporation

Land and Housing Development Corporation

Peoples HomeSite and Housing Corporation

Government Service Insurance System

Social Security System

Philippine Standards Association

A. R, Flores and Associates

National Building Code Committee

National Society for Seismology and Earthquake Engineering of the Philippines

Association of Structural Engineers of the Philippines

Philippine Institute of Civil Engineers

Bureau of Public Works

Philippine Civil Aeronautics Administration

National Science Development Board

USAID Mission to the Philippines

Bangladesh University of Engineering and Technology

CARE, Inc. - Bangladesh

Douet, Brown, Adams and Associates, Jamaica

U.S. National Bureau of Standards

In addition to NBS and AID support, principal members from the above groups donated their

time and professional expertise, test buildings, land to construct test buildings, scientific facilities and funding for continuation of Philippine research' activities

.

In mid-1973 two individuals were selected, representing the Bay of Bengal and the

northern Caribbean Island. These individuals contributed information about their respective geographic areas as inputs to the development of final design criteria. They will also

transfer the project results to their respective geographic areas.

Wind Research Activities

Field Test Sites

Three field test sites are being operated under the full-scale phase of the test program. In order of selection and installation of equipment, they are as follows: (1) Science

Garden at Quezon City, (2) Daet and (3) Laoag City. A number of factors were considered in

selecting these test sites; the final choices required some compromise in the original

selection criteria. The main factors in order of importance were as follows;

High frequency of extreme winds,

Accessibility

,

Availabiltiy of commercial power.

Type of wind exposure, and

Security
The Science Garden site was an obvious choice since it is staffed with qualified PAGASA

technicians. These technicians form the nucleus of a team trained by NBS in the installation, operation and maintenance of test equipment. It was thus possible for the NBS team

members to transfer a significant portion of the installation work and associated responsibilities during the early stages of the program. Science Garden also serves, in conjunction

with the University of the Philippines, as the center for reference standards, spare parts

and service equipment.

Selection of the remaining sites was not so obvious. PAGASA weather stations were

selected because of the security and personnel available to service the equipment. In

addition, measurements of temperature and barometric pressure would be available, thus freeing data channels for additional pressure measurements. In discussions with the Philippine

Advisory Committee, serious consideration was given to sites at Virac, Casiguran, Aparii,

Baguio City, Daet and Laoag City. Virac and Baguio City were eliminated on the basis of

unusual terrain features (mountain top locations) which are not typical of wind exposures

for housing developments and would be extremely difficult to model in the wind tunnel.

Casiguran was eliminated because of the difficulties of transporting equipment and personnel

for periodic maintenance and calibration (boat service only) . Aparii was eliminated because

of plans to relocate the present PAGASA weather station site, and the lack of electrical

power

.

Daet and Laoag City were selected with some compromise on statistical independence and

frequency of extreme winds. However, it is believed that other attributes of these sites

(ease of transporting equipment and personnel, availability of electric power, non-unusual

terrain features) more than offset these compromises.

Arrangements were made with the Philippine Advisory Committee to build test houses at

the sites and, where possible or necessary, to use existing PAGASA structures. Test equipment was assembled, tested, calibrated and packaged at the NBS for shipment to the test

sites

.

Science Garden, Quezon City . The wind exposure at this site varies with wind directicn,

being relatively clear and flat from N to SE and slightly rough from SE to W due to construction of high-rise buildings some 500 metres away. The site is moderately rough from

W to N due to local topographical features and other low-rise buildings on the test site.

Three test houses are instrumented. Basic plan dimensions of the first two gable roof units

are 7x8 metres. The third unit designed by CAKE, Inc. in Bangladesh was shipped to Quezon

City for full-scale testing. This building is barrel vaulted in end elevation and has plan

dimensions of 5.4 x 2.5 metres. The data acquisition system is located in the PAGASA

Instrumentation Building (adjacent to the first test house) which is air conditioned and

relatively free from dust. A total of 21 pressure channels are available (an increase of

10 channels through use of an auxiliary signal conditioning system)

.

Dust Weather Station . The Daet site is approximately 230 km SE of Manila. It is located next to the city airport which borders the Pacific Ocean. The site has a very flat and

clear exposure with the exception of a coconut grove running from NE to SE adjacent to the

ocean. The main PAGASA Station Building was instrumented. This is the only building under

X-4
test which has a hip roof. The data acquisition system is installed in the main building's

radar equipment room which is air conditioned. An emergency generating system is available

at this site in the event that commercial power service is disrupted.

Laoag City Weather Station . This site is approximately 500 km north of Manila. As

with Daet, the site is adjacent to the municipal airport and has a clear and flat exposure

for all directions. Two test houses with plan dimensions identical to the test units at

Science Garden (but with different roof slopes and eaves overhang) were constructed, one on

the PAGASA station ground and one on CAA property. One of the two units is located in

close priximity to two existing buildings which will allow the influence of neighboring

structures to be evaluated.

Research Equipment

Full-Scale Test Equipment . Based on technical needs, some of the data acquisition

equipment was specially developed at NBS while the balance was obtained from commercial

sources. The data acquisition system used in the full-scale test program consist of five

basic subsystems; (1) the sensors of transducers, (2) a logic section, (3) a signal conditioner, (4) a recorder, and (5) a power supply. The equipment was assembled, tested and

packaged for shipment to the Philippines by the NBS project staff.

The system is designed to continuously monitor input signals and to go into a calibration and record sequence when the signal being monitored exceeds a preset level. When the

recording period ends, the system enters a "hold" period during which no data are collected,

regardless of the signal level on the channel being monitored. The "record" and "hold"

periods are switch-selectable and are usually set at 20 and 30 minutes, respectively. The

total recording time available on a reel of tape is approximately 6 hours. Thus, the total

time period between changes in tape reels (assuming continuous high wind conditions) is

approximately 12 hours which is considered to be sufficient for most typhoon passages.

Wind speed and direction are measured by a propeller-vane anemometer mounted on a IOmeter mast located far enough from the test buildings to register conditions in the undisturbed wind field. The anemometer is rated at 100 m/s and provides the signal which triggers the recording system. An ambient pressure probe is also mounted on the mast just below

the anemometer and provides the differential pressure transducer with a standard reference

pressure. The pressure transducers are mounted in low-profile housings which are designed

to create a pressure intensity at their center equal to the pressure that would exist on the

surface of the building without the housings installed. This obviates the problems involved

with mounting the transducers flush with the wall or roof surfaces. The transducers are

fitted with a solenoid valve which is activated by the logic system one minute before the

generation of pressure records begins. This value places the transducer in a "closed- loop"

configuration, thus allowing the subsequent record to be corrected for zero offset due to

drift or loading of the transducer diaphragm with rainwater.

The logic portion of the data acquisition system normally operates in an automatic

mode, but provisions exist for manual intervention for the purpose of calibration or periodic system checkout. The logic section also includes a time code generator which provides

both digital and analog code of Greenwich Meridian Time (GMT) in days, hours, minutes and

X-5
seconds. The time code generator is set by radio countdown from PAGASA Headquarters in

Quezon City and normally requires correction only two or three times a year.

Analog signals from the pressure transducers are converted to DC voltages (demodulation) and filtered by the signal conditioner prior to recording. The signals are also

attenuated to match the range of the tape recorder, thus providing a better signal to noise

ratio than would be possible by straight recording of the transducer outputs.

The recordings section consists of a 14-track analog tape unit with 26.7 cm reels

containing 1100 metres of tape. Usual record speed is 4.8 cm/sec. The recorder is normally

in a "powered down" mode and only operates on command from the logic section. As with the

logic section, manual intervention is permitted for calibration, cleaning and rewinding tape.

Recorded signals can be reproduced in the field to check recorder operation.

With the exception of the Laoag City test site, commercial power is used to operate the

data acquisition systems. The Laoag City site is provided power by the CAA. To ensure availability of power under storm conditions, all three sites are equipped with a backup

system of batteries. The batteries are continuously charged when external power is

available and provide system power (115VAC) by means of an inverter. When commercial power

is interrupted, the batteries automatically switch on and pick up the load to supply the

data acquisition system for approximately 8 hours of continuous operation. The batteries

are recharged automatically when service is restored.

Since typhoon winds can be expected to come from any direction, it is extremely difficult to determine a "best" configuration of pressure transducers. Because roof structures

are known to be the most susceptible to wind damage, they received the highest priority in

transducer allocations. Extreme pressures acting along ridge lines, eaves and roof corners

are of interest as well as the average uplift pressures acting on the overall roof areas.

The configuration of pressure transducers is arranged differently on each test building,

thus providing the ability to measure a greater range of wind loadings. In all of the test

buildings, transducers were installed inside the building to measure internal pressures

which significantly influence the net roof uplift loads. An advantage of the test equipment used in this study is that the pressure transducer positions can be easily changed

after a storm to study wind pressure distributions on other parts of the test houses, thereby enhancing the value of the data and reducing the amount of redundant information.

Wind Tunnel Test Equipment . Test equipment being used in the wind tunnel can be divided

into three categories; (1) pressure transducers, (2) wind speed measuring equipment, and

(3) signal conditioning and analysis equipment. The pressure transducers are quite similar

to those being used in the full-scale tests except for the cylindrical transducer housings.

Nonnally, four transducers are installed in the building model although six units are

available

.

Wind speeds are measured by pitot tubes when mean values are required. For measurement of wind speed fluctuations, hot-wire anemometers are used. The anemometers have an

extremely high frequency response and are well suited to making measurement over the

frequency range of interest (0 to 500 Hz) . Two anemometer systems are used, permitting

simultaneous measurements at two points in the tunnel for the detemination of integral

X-6
scales (size of wind gusts) . The second system also provides backup if a component should

fail.

Wind Tunnel Modeling

The wind tunnel facility at the National Hydraulic Research Center (NHRC) , University

of the Philippines, is being used to carry out a series of tests on models of the full-scale

test buildings. The cross section of this tunnel is 1.22 metres square and 3.70 metres

long and produces a wind speed of approximately 30 metres per second. These model tests

have aided in the interpretation of full-scale studies and have allowed design pressure

coefficients to be determined in a systematic manner. While it is not possible to exactly model atmospheric boundary layers in conventional

wind tunnels, an acceptable degree of similitude is achieved by proper use of surface

roughness elements and vortex generators. This modeling technique used in the NHRC tunnel was used in previous wind tunnel model studies and was perfected for this applicaition in a slightly larger but quite similar wind tunnel at the Colorado State University (CSU)

.

Tapered spires are placed at the entrance to the test section to produce a sheared flow of

arbitrary turbulence intensity. The spires are followed by several rows of roughness elements located on the floor of the wind tunnel. The roughness elements generate a turbulent

boundary layer which extends to almost the full height of the tunnel at the downstream end

of the test section. Two combinations of spires and surface roughness elements were developed which produce turbulent boundary layers typical of smooth and moderately rough terrain,

respectively. The model scale being used in these studies is 1:80. Ideally, this scale

should be dictated by the integral scale of the turbulence, the effective surface roughness

height and the length associated with the peak of the turbulent energy spectrum. Test

results obtained in the CSU studies suggest a scale ratio of from 1:100 to 1:200. However,

the physical size of the model precludes installation of pressure transducers at these

small scales. Thus a compromise of 1:80 was used. Preliminary test results and results

obtained in other wind tunnel model investigations suggest that scale matching is of secondary importance compared to intensity of turbulence and shape of the spectral density

function.

Approximately 10 model configurations were investigated with roof slopes ranging from

0° to 30° and eaves overhangs ranging from 0 to 1.5 meters (full-scale). Two classes of

surface roughness were used in these studies. Several low-rise building types were analyzed

(houses, school buildings)

.

In addition to the wind tunnel modeling as described above, the NBS developed, specified, obtained and transported various items of test equipment for the UP wind tunnel

facility. The equipment was installed in the wind tunnel and training sessions were conduted in the operation of equipment and the interpretation of test results. Major items of

equipment transferred to the University of the Philippines include a signal correlator and

probability analyzer, hot-wire anemometers, pressure transducers, electronic filters,

signal amplifiers, x-y and stripchart recorders, voltmeters and an oscilloscope. This test

equipment is housed in an air conditioned room built by the University. Computer programs

X-7
developed at the NBS for the anlaysis of random data were transferred to UP Computer Center.

In addition, an extensive collection of documents dealing with the wind tunnel modeling of

buildings and other engineering structures was placed in the UP Library.

Collection, Reduction and Analysis of Field Data

All data collected at the field test sites are recorded on analog magnetic tapes.

Typically, these tapes contain one channel each of wind speed and direction data, 11 channels of pressure data and one channel of time code. Prior to the recording of data, a tape

is assigned a site designation and tape identification number. This information, along with

the time code, uniquely identifies the data.

The first step in the data reduction process is to record certain key channels , such

as wind speed, wind direction and a representative pressure signal on a paper stripchart

and subjectively classify the records by degree of stationarity . Records which contain

redundant information are eliminated at this time. The remaining records are then viewed on

an oscilloscope to determine the approximate maximum or minimum peak values and to verify

that the recordings have an acceptable signal to noise ratio and are free of discontinuities.

Once the records are determined to be of acceptable quality, they are converted to digital

form for analysis.

Analog to digital conversion is accomplished by means of a computer-controlled data

acquisition system which scans the analog channels in sequence and converts the voltage

levels into binary equivalents. The data channels are multiplexed at a rate of 20,000

channels per second so that the time skew is negligible for the frequency range of interest.

The scan rate can be varied but is typically 12 scans per second. Each channel is sampled

12,000 times, resulting in a record length of 1000 seconds (16 min - 40 sec). The multiplexing stage is followed by a programmable amplifier which allows best use of the digital

representation (eleven binary bits plus sign) . The digital data are entered on a 7-track

magnetic tape for subsequent analysis in the NBS Computer Center. This tape also contains

header information such as site and tape identification, the time of day when the original

data were recorded, and the length of record.

Several programs were developed at the NBS for the analysis of random data. These

include; Probability Density Function (PDF) which determines the peak values (either maximum or minimum) between zero crossings, calculates the mean and root mean square values

and plots probability distribution functions; Correlation Analysis (CORREL) which calculates correlation functions and spectral density estimates; and a Summation Program (SUMP)

which calculated the area-averaged surface pressures and the drag and uplift forces acting

on a structure.

Data obtained from the wind tunnel are processed "on line" and are not recorded for

future reduction or analysis. A hybrid computer allows the direct calculation of autoand cross-correlation functions as well as probability density and distribution functions.

A time domain analyzer is used to obtain direct measurements of mean and rms values. This

system has the disadvnatage of manual calculation of pressure coefficients, but this is

insignificant when compared with the ability to quickly assess the test results and alter

the model configuration without waiting for results from a central computer.

X-8,
Development of Design Pressure Coefficients

The 1970 edition of the National Building Code of Canada (NBC) provides for risk of

occurrence, terrain roughness, height above ground and building geometry in calculating

wind pressures,

P = qC C C . e g p

In this expression, q is a reference mean velocity pressure for a given mean recurrence

interval, C^ is an exposure factor which varies with surface roughness and height above

ground, C is a gust effect factor to provide for surface pressure fluctuations caused by

y turbulence and localized flow phenomena, and C^ is a conventional mean pressure coefficient.

The proper values for C^ will be determined from existing data, full scale wind data and

theoretical models of wind speed distributions in typhoons and hurricanes. The coefficients

Cg and C^ must be determined experimentally. This is the primary output of the wind tunnel

test program with the full-scale test results serving as a control on both the coefficients

and final design criteria.

Assessment, Selection and Application of Climatological Data

A study of the available Philippine wind climate information with a view of assessing

its adequacy from a structural engineering viewpoint will continue throughout the life of

the project. Records of wind speeds, of typhoon observations and damages due to significant typhoons were collected by NBS through the courtesy of the PAGASA. Out of these records, data were selected which appear to be suitable for analysis. These will be used as input in computer programs available at NBS for predicting extreme winds corresponding to

various mean recurrence intervals. A listing of the NBS programs was sent to the PAGASA

Computer Center for adoption by the PAGASA as a calculation tool. NBS will provide

appropriate assistance as required.

A parallel study of wind distributions, including the 1972 National Structural Code '

for Buildings of the Association of Structural Engineers of the Philippines and tropical

cyclone frequency and intensity maps, reveals the area of northern Luzon, and perhaps parts

of Western Luzon (including Manila) may need to be included in a more intense wind zone

area. It is desirable that future editions of the building code differentiate between

zones with different exposure (e.g. urban terrain vs. coastal sites). Results from this

task will provide the appropriate Philippine code officials with much needed information

for incorporation in a new building code. This information will serve as an input to the

final report.

Information Transfer

In addition to advisory committee meetings, conferences and workshops provide a mechanism to transfer information to a larger body of individuals.

An International Workshop held in Manila, Philippines on November 14-17, 1973 addressed

the state-of-the-art in mitigating building damages from winds. The workshop was jointly

sponsored by AID, the Philippine Advisory Committee and the U.S. National Bureau of Standaids.

X-9
Four themes covering climatology and aerodynamics, structural engineering, socio-economic and architectural considerations, and codes and standards were discussed.

The first two workshop days were devoted largely to presentations of technical papers.

During the afternoon of the first day, a visit was scheduled to the field test site at the

PAGASA Science Garden site, Quezon City.

Nine papers and five related reports were presented during the technical sessions with

time reserved for discussions. The third day was devoted to subcommittee working sessions

where 31 recommendations were developed. They were presented and discussed on the fourth

day. Approximately 140 individuals from five countries (Jamaica, Bangladesh, the United

Kingdom, the Philippines and the United States) attended the workshop.

The proceedings of the workshop was published as an NBS Building Science Series 56.

This publication includes 14 recommendations (edited from 31) for improved building practices, the opening ceremonies, and the technical papers and reports.

On May 16-17, 1975 a regional conference was conducted in Manila, Philippines to

discuss the draft project results to date. This meeting afforded the opportunity for mem- bers from wind-prone countries to further establish a dialogue for suggesting methods to

better present the final results. A similar regional conference will be conducted in

Kingston, Jamaica on November 6-7, 1975. The final report is expected to be published by

the end of June 1975.

Conclusion

This paper descirbes the activities associated with developing improved design criteria

for low-rise building to better resist the effects of extreme winds. It also discusses a

very important element associated with developing technology—its transfer to the ultimate

user. After completing the research a continuing effort by the Philippine Advisory

Committee will be required to further the development of technology. This project is only

one step in the process of improvement that will continue.

Acknowledgments

Parts of this paper were excerpted from NBSIR 74-567 FY 74 Progress Report on Design

Criteria and Methodology for Construction of Low-Rise Buildings to Better Resist Typhoons

and Hurricanes by Raufaste, N.S. and Marshall, R.D., Center for Building Technology, National Bureau of Standards, Washington, D.C. 20234.

X-10 i
References

1. Climatology and Wind Related Problems in Philippines" by R.L. Kintanar, November 1973, Manila, Philippines (paper presented at a workshop in Manila, Philippines)

.

X-il
SURVEY ON SEISMOLOGY AND EARTHQUAKE

ENGINEERING IN INDIA, IRAN AND TURKEY

by

Makoto Watabe

International Institute of Seismology

and Earthquake Engineering

and

Hideo Tokuhiro

Ministry of Construction

and

Masakazu Shinozuka

Japan International Cooperation Agency

ABSTRACT

A series of earthquake engineering courses have been held in Japan in cooperation with

Iran, India and Turkey. This paper presents the details of these courses and the aid such courses have provided to the supporting covmtries.

Key Words: Earthquake engineering; Earthquakes; Education; India; Iran; Training; Turkey.

X-12
Summary

1) A training course was conducted at the International Earthquake Engineering Department of Building Research Institute, Ministry of Construction, which is greatly appreciated.

2) It is required to establish an advanced course for these ex-participants in order

to provide an up-to-date knowledge and technique on earthquake engineering. In this con- nection, a 3 to 6 months' training course, is being organized.

3) Recrutiment procedures for the training course should be made promptly (4 to 6

months prior to the commencement of the training course)

.

4) In regard to the follow-up of the ex-participants, it is necessary for a staff

from either the Japanese Embassy or the JICA overseas office to keep close contact with

their representatives. Such a successful case concerning this plan has been Turkey.

5) In addition to inviting a technical trainee, it is important to promote a visit

from Japanese expert to the trainees country.

6) It is the intention of this program to strengthen our system so as to cope with

the request from ex-participants, and to readily supply technical information and

measuring instruments. Furthermore, an invitation of high ranking officials should be considered at the earliest date, in order to promote further technical cooperation in this

field.

7) A follow-up team was well received in every country. It is required to accelerate

this kind of service more positively, but more carefully in the future. If this follow up

team is dispatched every three years and the selection of the participants for the advanced

course is made at that time, the intensive training in Japan will be more beneficial.

Analysis of Questionnaire

1) Ex-participants' position in their countries are as follows; 86% of the ex-participants belong to the public or educational institution, and two thirds of them hold an

important post. From this trend, it can be easily understood that their training in Japan

was highly regarded and they were promising and selected researchers.

2) Evaluation of training

60% of the ex-participants or their senior officials highly praised their training

in Japan.

3) Composition of the personnel at I.I.S.E.E. (International Institute of Seismology

and Earthquake Engineering)

:

71% of the ex-participants are in favor of the present composition of our Insititute,

while 15% of them expect an increase in the number of staff, and 4% hope that it will be

enlarged. 19% of the ex-participants strongly wish to introduce lecturers from abroad, and

57% agree, while 24% think it necessary, or a matter of no consequence. As a conclusion,

it can be interpreted that many of the ex-participants expect to have foreign lecturers

added to the staff of the Institute.

4) Training System:

Each ex-participant enthusiastically hoped to be re-trained in a short period during

the training course. In connection with the duration of training, 62% of the ex-particiX-13
pants hoped it would be less than 6 months, while 14% of them wanted a longer period.

During this training, it is requested that more attention to the selection of applicants

be paid, in order to allocate more time for special self-study, laboratory practice and

an exchange of views on seismology and earthquake engineering.

5) Future follow-up service;

We have encountered numerous inquiries for catalogues of Japan-made measuring instruments and information on how to purchase these and if there is a sales agent in their

country. Many of these engineers wanted a continuous supply of the research reports and the

"Year Book" regularly published by the Institute, or the periodical "Technocrat". Furthermore, as the best means of follow-up, some engineers eagerly requested that Japanese ex- perts be sent to their country.

Situation of Seismology and Earthquake Engineering in India, Iran and Turkey

1) India

The Japanese Ambassador has asked whether there are earthquakes or not in India.

As such an event seldom occurs, people are quite indifferent to earthquakes. For that

reason, it seems that the officials in charge of the technical cooperation at the Ministry

of Finance in India will take a negative attitude toward allocating scholarships offered

from Japan to this training course , although it is recognized that in the Meteorological

Department of Roorkee University where there are many young researchers eager to study in

England, United States, or Japan. Although the directors at the Meteorological Department

have intended to send at least one researcher to our Institute every year, only one has come

in the past 2 or 3 years to study in Japan. Due to the influence of England, seismology in

India has advanced considerably as has earthquake engineering. In Roorkee University,

located in about 200km north of New Delhi, Dr. J. Krishna and his excellent staff have

established the Earthquake Engineering Center, and have invited young researchers from

neighboring countries to study. However, Japan's follow up service has been badly evaluated in this country. For example, in regard to Japanese made measuring instruments, due

to imperfection in the catalogue and specifications it is difficult to repair these instruments when they are out of order. Accordingly, the donation of experimental instruments

should be made providing a service engineer be made available. On the other hand, the United

States, because of the demand made when the instruments are out of order, replaced in a

week through the diplomatic channels. Also relative to follow up service, Japan is far

behind the United States

.

2) Iran

It seems that Iran, favored with abundant oil, is seeking a well-balanced diplomatic

policy toward the big powers and in part in the technical cooperation. Iranian people are

in fear of earthquakes and thus the Government of Iran has given a considerable thought

to this matter. In spite of active influence by the Ministry of Housing Development, a

very limited number of participants have been conveyed to the Japanese Embassy, due to a

difference in views with other organizations concerned with technical cooperation.

Recently, a Building Research Center has been established by a joint project with UNESCO,

X-14

1
in Iran. Mr. Javdan, one of the ex-participants, has been engaged as director of this

research work. His contribution toward the establishment of this organization is to be

highly praised. This institution was formed for the purpose of providing leadership in the

field of the earthquake engineering in the Middle and Near East. They disclosed a desire

to purchase research equipment from Japan, nevertheless Mr. Pakdaman, Vice-Minister of

the Housing Development expressed his regret to Japan because it took a long time and re- quired sophisticated formalities in order to purchase such items. However, a feeling of

satisfaction was obtained to see a dome-shaped housing complex with aseismatic structure,

designed by the ex-participants which was constructed in the suburb of Gazvin city about

150km west of Tehran.

It seems that an ideal type of technical cooperation between Japan and Turkey can be

obtained. Ex-participants who studied in Japan are now taking initiative in this field

and in particular research work relative to earthquake engineering. In addition, a close

liaison among the ex-participants, both in the Earthquake Engineering and in Seismology,

has been preserved at the same time, and there is constant contact with the Japanese Embassy. Success in this technical cooperation greatly owed to the prominent Japanese experts

who made self-sacrificing efforts when in Turkey. It is assured that Japanese technology

in the area of earthquake engineering and seismology will further help advance Turkey's

technology

.

Special Lecture

A follow up team made an attempt to present a special lecture on "Evaluation of

Seismic Input Force" through a 16mm film of the earthquake engineering. To facilitate

the understanding of the lecture, a series of color slides, and moreover a lecture notes

were distributed to attendents. After 45 minutes of lecture, a short motion picture

entitled "Vibration" was then shown, and a special lecture was then concluded with a question period. Because the actual lecture hall was not properly selected, some difficulty

arose with respect to presentation of slides. However, the lectures were a success due to

the kindness of the Japanese Embassy and the JICA overseas office. The special lecture

was well received by attendants. The number of attendance in the respective countries

were as follows;

3) Turkey

India 160

Iran 70

Turkey 70

Total 300

Contents of the lecture

iv)

i)

ii)

iii)

Stochastics on the history of earthquakes

Magnitude and intensity and the maximum acceleration

Evaluation of the intensity level of the earthquake in terms of return period

The earthquake damage and subsoil condition

X-:15
v) Aseismic regulations and response spectriim

vi) International aseismic regulations

As for movie films

i) Ship and wave

ii) Earthquake and structure

iii) Wind and suspension bridge

iv) Air turbulance and air plane

v) Road and automobile

vi) Derailing and vibration of the train

vii) Utilization of vibration

* U. S. GOVERNMENT PRINTING OFFICE : 1977 O - 231-282

X-16
<;yfBS-M4A (REV. 7-73) - ^

U.S. DEPT. OF COMM. 1. PUBLICATION OR REPORT NO. 2. Gov't Accession

SHEET ^"^^^

3. Recipient's Accession No.

4. TITLE AND SUBTITLE „^ WIND AND SEISMIC EFFECTS

( PROCEEDINGS OF THE SEVEiSITH JOINT PANEL CONFERENCE OF THE

U.S. -JAPAN COOPERATIVE PROGRAM IN NATURAL RESOURCES N

5. Publication Date

April 1977

6. Performing Organization Code

/• A U 1 MUK^o;

H. S. Lew, Editor

8. Performing Organ. Report No.

9. PERFORMING ORGANIZATION NAME AND ADDRESS

NATIONAL BUREAU OF STANDARDS

DEPARTMENT OF COMMERCE

10. Project/Task/Work Unit No.

11. Contract/Grant No.

12. Sponsoring Organization Name and Complete Address (Street, City, State, ZIP)

Same as Item # 9.

13. Type of Report & Period

Covered

14. Sponsoring Agency Code

15. SUPPLEMENTARY NOTES

library of Congress Catalog Card Number: 77-608015

16. ABSTRACT (A 200-word or less factual summary ol most sigrtificant information. If document includes a significant

bibliography or literature survey, mention it here.)

The Seventh Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects was

held in Tokyo, Japan on May 20-23, 1975. The proceedings of the Joint Meeting include

the program, the formal resolutions, and the technical papers. The subject matter

covered in the DaoerS incllldPS Ph^Tafl-pTl «;"|-'ir'C! o-F Q-t-mnrr wi nr1« racr^r^ca r\-f -Fit 1 1 _o/-i=j 1

Q

Structures to wind action; geological distribution of seismic activity; maintenance

of strong motion accelerographs and data processing; strong earthquake notions and

ground failures; response of hydraulic and building structures to seismic forces;

aseisraic considerations for vessels; recent revisions of design standards on wind

and seismic effects; joint research program utilizing large scale testing facilities;

and technological assistance to developing countries.

17. KEY WORDS (six to twelve entries; alphabetical order; capitalize only the first letter of the first key word unless a proper name; separated by semicolons) Accelerograph; bridges; buildings; codes; disaster;

dynamic analysis; earthquakes; ground failures; hydraulic structures; seismicity;

soils; standards; structural response; vessels and wind.

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PERIODICALS

JOURNAL OF RESEARCH reports National Bureau

f Standards research and development in physics,

lathematics, and chemistry." It is published in two

actions, available separately:

Physics and Chemistry (Section A)

apers of interest primarily to scientists working in

lese fields. This section covers a broad range of physiil and chemical research, with major emphasis on

;andards of physical measurement, fundamental con-

;ants, and properties of matter. Issued six times a year,

nnual subscription: Domestic, $17.00; Foreign, $21.25.

Mathematical Sciences (Section B)

tudies and compilations designed mainly for the mathnatician and theoretical physicist. Topics in mathematal statistics, theory of experiment design, numerical

nalysis, theoretical physics and chemistry, logical deign and programming of computers and computer sys- ms. Short numerical tables. Issued quarterly. Annual

ubscription: Domestic, $9.00; Foreign, $11.25.

! DIMENSIONS/NBS (formerly Technical News Bullen)—This monthly magazine is published to inform

i^ientists, engineers, businessmen, industry, teachers,

;udents, and consumers of the latest advances in

!:ience and technology, with primary emphasis on the

'ork at NBS. The magazine highlights and reviews

|ich issues as energy research, fire protection, building

;chnology, metric conversion, pollution abatement,

3alth and safety, and consumer product performance.

Ii addition, it reports the results of Bureau programs

ji measurement standards and techniques, properties of

|.atter and materials, engineering standards and serves, instrumentation, and automatic data processing.

Annual subscription: Domestic, $12.50; Foreign, $15.65.

I NONPERIODICALS

onographs—Major contributions to the technical liter- ,ure on various subjects related to the Bureau's scienfic and technical activities.

'andbooks—Recommended codes of engineering and

idustrial practice (including safety codes) developed

I cooperation with interested industries, professional

•ganizations, and regulatory bodies.

)ecial Publications—Include proceedings of conferences

l)onsored by NBS, NBS annual reports, and other

oecial publications appropriate to this grouping such

) wall charts, pocket cards, and bibliographies,

pplied Mathematics Series—Mathematical tables, manils, and studies of special interest to physicists, engiBers, chemists, biologists, mathematicians, com-

(jter programmers, and others engaged in scientific

fid technical work.

ational Standard Reference Data Series—Provides

aantitative data on the physical and chemical proper-

!es of materials, compiled from the world's literature

fid critically evaluated. Developed under a world-wide

rogram coordinated by NBS. Program under authority

National Standard Data Act (Public Law 90-396).

NOTE: At present the principal publication outlet for

these data is the Journal of Physical and Chemical

Reference Data (JPCRD) published quarterly for NBS

by the American Chemical Society (ACS) and the American Institute of Physics (AIP). Subscriptions, reprints,

and supplements available from ACS, 1155 Sixteenth

St. N.W., Wash. D. C. 20056.

Building Science Series—Disseminates technical information developed at the Bureau on building materials,

components, systems, and whole structures. The series

presents research results, test methods, and performance criteria related to the structural and environmental

functions and the durability and safety characteristics

of building elements and systems.

Technical Notes—Studies or reports which are complete

in themselves but restrictive in their treatment of a

subject. Analogous to monographs but not so comprehensive in scope or definitive in treatment of the subject area. Often serve as a vehicle for final reports of

work performed at NBS under the sponsorship of other

government agencies.

Voluntary Product Standards—Developed under procedures published by the Department of Commerce in Part

10, Title 15, of the Code of Federal Regulations. The

purpose of the standards is to establish nationally rec- ognized requirements for products, and to provide all concerned interests with a basis for common understanding of the characteristics of the products. NBS

administers this program as a supplement to the activities of the private sector standardizing organizations.

Consumer Information Series—Practical information,

based on NBS research and experience, covering areas

of interest to the consumer. Easily understandable language and illustrations provide useful background knowledge for shopping in today's technological marketplace.

Order above NBS publications from: Superintendent

of Documents, Government Printing Office, Washington,

D.C. 20i02.

Order following NBS jmblications—NBSIR's and FIPS

from the National Technical Information Services,

Springfield, Va. 22161.

Federal Information Processing Standards Publications

(FIPS PUBS)—Publications in this series collectively

constitute the Federal Information Processing Standards Register. Register serves as the official source of

information in the Federal Government regarding standards issued by NBS pursuant to the Federal Property

and Administrative Services Act of 1949 as amended.

Public Law 89-306 (79 Stat. 1127), and as implemented

by Executive Order 11717 (38 FR 12315, dated May 11,

1973) and Part 6 of Title 15 CFR (Code of Federal

Regulations).

NBS Interagency Reports (NBSIR)—A special series of

interim or final reports on work performed by NBS for

outside sponsors (both government and non-government). In general, initial distribution is handled by the

sponsor; public distribution is by the National Technical Information Services (Springfield, Va. 22161) in

paper copy or microfiche form.

BIBLIOGRAPHIC SUBSCRIPTION SERVICES

Superconducting Devices and Materials. A literature

survey issued quarterly. Annual subscription: $20.00.

Send subscription orders and remittances for the preceding bibliographic services to National Bureau of

Standards, Cryogenic Data Center (275.02) Boulder,

Colorado 80302.

he following current-awareness and literature-survey

libliographies are issued periodically by the Bureau:

ryogenic Data Center Current Awareness Service. A

literature survey issued biweekly. Annual subscription: Domestic, $20.00; Foreign, $25.00.

iquified Natural Gas. A literature survey issued quarisrly. Annual subscription: $20.00.
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