A 11 ID 5 DTflSTb

NIST

PUBLICATIONS

NBS SPECIAL PUBLICATION 665

U.S. DEPARTMENT OF COMMERCE/National Bureau of Standards

DC

Wind and Seismic

Effects

Proceedings of the

Twelfth Joint UJNR

Panel Conference

100

.157

Mo, 665

1984
NATIONAL BUREAU OF STANDARDS

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the Institute for Computer Sciences and Technology.

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1 Headquarters and Laboratories at Gaithersburg, M D, unless otherwise noted; mailing address Washington, DC 20234. 'Some divisions within the center are located at Boulder, CO 80303.
Wind and Seismic Effects

Proceedings of the 12th Joint

Panel Conference of the U.S.-Japan

Cooperative Program in

Natural Resources

May 19-23, 1980

National Bureau of Standards

Gaithersburg, MD

E.V. Leyendecker and R.M. Chung, Editors

Center for Building Technology

National Engineering Laboratory

National Bureau of Standards

Washington, DC 20234

U.S. DEPARTMENT OF COMMERCE, Malcolm Baldrige, Secretary

NATIONAL BUREAU OF STANDARDS, Ernest Ambler, Director

Issued January 1 984

to

Q
Library of Congress Catalog Card Number: 83-600593

National Bureau of Standards Special Publication 665

Natl. Bur. Stand. (U.S.), Spec. Publ. 665, 543 pages (Jan. 1984)

CODEN: XNBSAV

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 1984

For sale by the Superintendent of Documents, U.S. Government Printing Office Washington, D.C. 20402
PREFACE

The Twelfth Joint Meeting of the U.S. - Japan Panel on Wind and

Seismic Effects was held in Gaithersburg, Maryland on May 19-23, 1980.

This panel is a part of the U.S. - Japanese Cooperative Program in Natural

Resources (UJNR) . The UJNR was established in 1964 by the U.S. - Japan

Cabinet-level Committee on Trade and Economic Affairs. The purpose of

the UJNR is to exchange scientific and technology information which will

be mutually beneficial to the economics and welfare of both countries.

These proceedings include the program, the formal resolutions, and

the technical papers presented at the Joint Meeting. The texts of the

papers have been edited for obvious errors and clarity.

It should be noted that throughout the proceedings certain commercial equipment, instruments or materials are identified in order to

specify adequately experimental procedure. In no case does such indentication imply recommendation or endorsement by the National Bureau of

Standards, nor does it imply that the material or equipment identified

is necessarily the best for the purpose.

Preparation of the proceedings was partially supported by funds

from the National Science Foundation.

H. S. Lew, Secretary

U.S. Panel on Wind and

Seismic Effects

Request # 52 of 74 . Screen 1 of 2

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ABSTRACT

The Twelfth Joint Meeting of the U.S. -Japan Panel on Wind and

Seismic Effects was held in Gaithersburg , Maryland on May 19-23, 1980.

The proceedings of the Joint Meeting include the program, the formal

resolutions, the Task Committee Reports, and the technical papers. The

subjects covered in the papers include: (1) the characterization of

seismic ground motion, (2) the characterization of natural wind and extreme

wind records, (3) structural response to earthquake loading, (4) storm

surge and tsunamis, (5) recent developments in seismic design criteria,

(6) technical cooperation with developing countries, (7) earthquake hazard

mitigation, and (8) structural response to wind loading.

KEY WORDS: accelerograph; bridges; codes; design criteria; disaster;

earthquakes; geotechnical engineering; ground failures; seismic

design; seismicity; standards; structural engineering;

tsunamis; winds.

iv
SI CONVERSION UNITS

In view of the present accepted practice for wind and seismic technology, common units of

measurements were used throughout this publication. In recognition of the position of the United

States as a signatory to the General Conference on Weights and Measures, which gave official status

to the International System of Units (SI) in 1960, the table below is presented to facilitate conversion to SI Units. Readers interested in making further use of the coherent system of SI units are

referred to: NBS SP 330, 1977 Edition, The International System of Units; and ASTM Standard for

Metric Practice.

TABLE OF CONVERSION FACTORS TO SI UNITS

Customary Units

International

(SI) UNIT

Conversion

Approximate

Length inch (in) meter (m) a

1 in = 0.0254 m*

foot (ft) meter (m) 1 ft = 0.3048 m*

Force pound (lbf) newton (N) 1 lbf = 4.48 N

kilogram (kgf) newton (N) 1 kgf = 9.807 N

Pressure or Stresspound per square

inch (psi)

kip per square

newton/per square

meter

newton per square

1 psi = 6895 N/m2

inch (ksi) meter 1 ksi = 6895 x 10 3 N/m2

Energy inch«pound (in-lbf) joule (J) 1 in-lbf = 0.1130 J

foot-pound (ft-lbf) joule (J) 1 ft-lbf = 1.3558 J

Torque

or Bending

Moment

pound«inch newton-meter (N«m) 1 lbf. in = 0.1130 N«m

pound»foot (lbf-ft) newton-meter (N«m) 1 lbf .ft = 1 .3558 N-m

Weight pound (lb) kilogram (kg) 1 lb = 0.4536 kg

Unit Weight pound per cubic foot

(pcf)

kilogram per cubic

meter (kg/m3

)

1 pcf = 16.018 kg/m3

Velocity foot per second

(ft/sec)

meter per second

(m/s)

1 ft/s = 0.3048 m/s

Acceleration foot per second

square (ft/s 2

) meter per second

square (m/s 2

)

1 ft/s 2 = 0.3048 m/s 2

a Meter may be subdivided. A centimeter (cm) is 1/100 m and a millimeter (mm) is 1/1000 m.

* Exactly

CONTENTS

Page

Preface iii

Abstract iv SI Conversion Units v Program of the Joint Meeting x

List of Members xiv

Resolution of the Twelfth Joint Meeting xix

Themes and Technical Papers

Theme I: CHARACTERIZATION OF SEISMIC GROUND MOTION

Dense Instrument Array Program of the Public Works Research Institute for

Observing Strong Earthquake Motion 1 Yoshiziro Sakagami , Tadayoshi Okubo, Toshio Iwasaki, Kazuhiko Kawashima

U.S. Geological Survey Strong-Motion Records from the Imperial Valley

Earthquake, October 15, 1979, Preliminary Summary 29

J. R. Filson, R. L. Porcella, R. B. Matthiesen

Three-Dimensional Synthetic Accelerograms for Design Earthquake Ground

Motions 40

Makoto Watabe

Study of Short and Long Period Dynamic Ground Characteristics in the Sendai

District 56 Yoshikazu Kitagawa, Makoto Watabe

Theme II: CHARACTERIZATION OF NATURAL WIND AND EXTREME WIND RECORDS

Hurricane Wind Speeds in the United States 67 Martin E. Batts, Larry R. Russell, Emil Simiu

A Statistical Result on Maximum Surface Winds Surrounding Typhoons 89 Shigemi Fujiwhara, I. Subbaramayya

Surface Winds from Hurricane Frederic: An Engineering Viewpoint 96 Timothy A. Reinhold

Hurricanes David and Frederic 1979 115

Celso S. Barrientos

Theme III: STRUCTURAL RESPONSE TO EARTHOUAKE LOADING

Seismic Response of Floating Offshore Structures 117 Hajime Tsuchida, Susumu Iai , Satsuo Noda

Critical Excitations of Structures 133

R. F. Drenick, F. Novomestky, G. Bagchi

Dynamic Testing of a Three-Story Reinforced Concrete Structure 143

Keiichi Ohtani

Buckling Analysis of Buried Pipelines under Seismic Loads 156

C. C. Chen, T. Ariman, L. H. N. Lee

Shaking Table Tests of Earth Retaining Structures on Improved Ground 167

Setsuo Noda, Sosuke Kitazawa, Hajime Tsuchida

via
Page

Full-Scale Experimental Study on the Aseismic Performance of Medium-Rise

Reinforced Concrete Wall Structures 187 Masaya Hirosawa, T. Goto, H. Hiraishi and M. Yoshimura

Soil Dynamics - State of the Art, 1980 213

William F. Marcuson III, A. G. Franklin, P. F. Hadala

Dynamic Analysis of Sheet Pile Foundations 235

Tatsuo Asama, Yukitake Shioi, Michio Okahara, Yasuo Mitsuie

Dynamic Behavior of Multi-column Foundations with Inclined Piles during

Earthquakes 243

Tatsuo Asama, Yukitaki Shioi, Yasuo Mitsuie

Analysis of Liquefaction Potential of Sands on Level Ground During Earthquakes

Using Cyclic Strain Controlled Tests 260

Felix Y. Yokel, Riley M. Chung, Ricardo Dobry and Richard S. Ladd

The Participation Factor of Horizontal Force Applied to Pile Foundations 271 Yashihiro Sugimura, Makoto Watabe

Theme IV: STORM SURGE AND TSUNAMIS

Operational Models for Forecasting Storm Surges in NOAA 297

Celso S. Barrientos

Storm Surge Prevention Around River Mouths in Japan 301

Hiroshi Hashimoto

Theme V: RECENT DEVELOPMENT IN SEISMIC DESIGN CRITERIA

New Structural Design Enforcement for Building Structures 316

Kiyoshi Nakano, Makoto Watabe, Yuji Ishiyama

Impact of Draft Highway Bridge Seismic Design Guidelines 328

James D. Cooper, Ronald L. Mayes, Roland L. Sharpe

New Specifications for Earthquake-Resistant Design of Japanese Highway

Bridges 341

Eiichi Kuribayashi , Toshio Iwasaki, Osamu Ueda

Single-Story Residential Masonry Construction in Uniform Building Code Seismic

Zone - Tentative Recommendations 365

Ronald Morony

Theme VI: TECHNICAL COOPERATION WITH DEVELOPING COUNTRIES

Earthquake Disaster Preparedness Program for the Lima, Peru Metropolitan

Area 378

Masamitsu Ohashi , Makoto Watabe, Yoshio Kumagai, Yoshihisa Hoshino

Fundamental Factors in Optimizing Earthquake Disaster Mitigation

Investments 398

Yoshijiro Sakagami , Eiichi Kuribayashi, Osamu Ueda and Tadayuki Tazaki

Wood Sheathed Diaphragms at Lateral Force Resisting Elements 409

Edwin G. Zacher

vii i
Page

A Proposed Method for the Evaluation of Aseismic Performance of Existing

Timber Dwellings 429

Tatsuro Murota , H. Okada

An Evaluation Method on the Aseisraicity of Earth Structures (Road

Embankments) 442

Tatsuya Fujii, Yasushi Sasaki

Wind Induced Vibration of Truss Members of the Commodore Barry Bridge 460

Gerald F. Fox

Calculation of the Gust Responses of Long-Span Bridges (II) 485

Nobuyuki Narita, Hiroshi Sato

Active Turbulence Generator for Aerodynamic Tests of Bridge Section Models ... 49 5 B. Bienkiewicz, R. J. Cermak, J. Peterka

APPENDIX A - TASK COMMITTEE REPORTS 506

ix
PROGRAM OF THE TWELFTH JOINT MEETING

OF THE

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS

MAY 19-22, 1980 AT THE NATIONAL BUREAU OF STANDARDS

GAITHERSBURG, MARYLAND

MONDAY - May 19 OPENING SESSION

10:00 a.m. Call to order by Dr. H. S. Lew, Secretary, U.S. Panel

Remarks by Dr. Thomas A. Dillon, Deputy Director, National Bureau of

Standards

Remarks by Mr. Hiroyoshi Kurihara, Counselor, Embassy of Japan

Remarks by Dr. Edward 0. Pfrang, Chairman, U.S. Panel

Remarks by Mr. Yoshijiro Sakagami, Chairman, Japanese Panel

10:30 a.m. Introduction of U.S. Panel Members by U.S. Chairman and Japan Panel Members

by Japanese Chairman

10:45 a.m. Election of Conference Chairman

10:50 a.m. Adoption of Agenda

1:00 p.m. Dense Instruments Array Program of the Public Works Research Institute for

Observing Strong Earthquake Motion — Y. Sakagami, T. Okubo, T. Iwasaki,

K. Kawashima

1:30 p.m. Recent Strong-Motion Results from Imperial Valley Earthquake, October 15,

1979 — J. R. Filson

2:00 p.m. Three-Dimensional Synthethic Accelerograms for Design Earthquake Ground

Motion — M. Watabe

2:20 p.m. Study of Short and Long Period Dynamic Ground Characteristics in the

Sendai District — M. Watabe, Y. Kitagawa

2:40 p.m. Discussion

3:20 p.m. Hurricane Wind Speeds in the United States — M. E. Batts, L. R. Russell,

E. Simiu

3:50 p.m. A Statistical Result on Maximum Surface Winds Surrounding Typhoons —

S. Fujiwhara, I. Subbaramayya

4:10 p.m. Surface Winds from Hurricane Frederic: An Engineering Viewpoint —

T. Reinhold

4:30 p.m. Hurricanes David and Frederic 1979 — C. Barrientos

4:50 p.m. Discussion

x
5:00 p.m. Task Committee Meetings

6:00

TUESDAY

9:00 a.m.

9:30 a.m.

10:00 a.m.

10:20 a.m.

10:40 a.m.

11:10 a.m.

11:30 a.m.

11:50 a.m.

1:00 p.m.

1:20 p.m.

1:40 p.m.

2:00 p.m.

2:20 p.m.

2:40 p.m.

3:20 p.m.

3:40 p.m.

4:00 p.m.

A. Strong-Motion Instrumentation Arrays and Data

B. Large-Scale Testing Program

C. Repair and Retrofit of Existing Structures

H. Soil Behavior and Stability during Earthquakes

Adjourn

May 20

Seismic Response of Floating Offshore Structures — H. Tsuchida, S. lai,

S. Noda

Critical Excitation of Structures — R. F. Drenick, G. Bagchi

Dynamic Test of Three-Story Reinforced Concrete Structure — K. Ohtani

Buckling Analysis of Buried Pipelines under Seismic Loads — C. C. Chen,

T. Ariman, L. H. N. Lee

Discussion

Shaking Table Tests of Earth Retaining Structures on Improved Grounds —

S. Noda, S. Kitazawa, H. Tsuchida

Full-Scale Experimental Study on Aseismic Performance of Medium-Rise

Reinforced Concrete Wall Structure — M. Hirosawa, T. Goto, H. Hiraishi,

and M. Yoshimura

Discussion

Soil Dynamics - State of the Art, 1980 — W. F. Marcuson III,

A. G. Franklin, P. F. Hadala

Dynamic Analysis of Sheet Pile Foundation — T. Asama, Y. Shioi,

M. Okahara, Y. Mitsuie

Dynamic Behavior of Multicolumn Foundation with Inclined Piles during

Earthquake — T. Asama, Y. Shioi, H. Asaruma, Y. Mitsuie

Analysis of Liquefaction Potential of Sands on Level Ground During

Earthquakes Using Cyclic Strain Controlled Tests — F. Y. Yokel,

R. M. Chung, R. Dobry, and R. S. Ladd

The Participation Factor of Horizontal Force Applied to Pile Foundations —

Y. Sugimura, M. Watabe

Discussion

Operational Model for Forecasting Storm Surges in NOAA — C. Barrientos

Storm Surge Prevention Around the River Mouths in Japan — H. Hashimoto

Discussion

xi
4:10 p.m. Task Committee Meetings

5:00 p.m.

WEDNESDAY

9:00 a.m.

9: 30 a.m.

10:00 a.m.

10:40 a.m.

11:00 a.m.

11:20 a.m.

11:50 a.m.

1: 30 p.m.

2:00 p.m.

2:30 p.m.

2: 50 p.m.

3:10 p.m.

3:50 p.m.

5:00 p.m.

THURSDAY

9:00 a.m.

9:30 a.m.

10:00 a.m.

10:30 a.m.

D. Structural Performance Evaluation

E. Land Use Programs for Controlling Natural Hazard Effects

F. Disaster Prevention Methods for Lifeline Systems

G. High Winds (Extreme Winds and Structural Response)

H. Soil Behavior and Stability During Earthquakes

I. Storm Surge and Tsunami

Adjourn

May 21

New Structural Design Enforcement for Building Structures — K. Nakano,

M. Watabe

Impact of Draft Highway Bridge Seismic Design Guidelines — J. D. Cooper,

R. L. Mayes, R. L. Sharpe

New Specifications for Earthquake-Resistant Design of Highway Bridges in

Japan — E. Kuribayashi, T. Iwasaki, 0. Ueda

Single-Story Residential Masonry Construction in Uniform Building Code

Seismic Zone - Tentative Recommendations — R. J. Morony

Discusion

Earthquake Disaster Preparedness Program of Lima, Peru — M. Watabe

Discussion

Fundamental Factors in Optimizing Earthquake Disaster Mitigation

Investment — Y. Sakagami , E. Kuribayashi, 0. Ueda, and T. Tazaki

Wood Sheathed Diaphragms as Lateral Force Resisting Elements — E. Zacher

A Proposed Method for the Evaluation of Aseisraic Performance of Existing

Timber Dwellings — T. Murota, H. Okada

An Evaluation Method on Aseismiclty of Earth Structures (Road

Embankments) — T. Fujii, Y. Sasaki

Discussion

Tour of The National Bureau of Standards

Adjourn

May 22

Wind-Induced Vibration of Truss Members of the Commodore Barry Bridge —

G. F. Fox

Calculation of the Gust Responses of Long-Span Bridges (II) — N. Narita,

H. Sato

Active Turbulence Generator for Wind Tunnel Study of Bridge Aerodynamic

Stability — R. J. Cermak, B. Bienkiewicz, J. Peterka

Discussion

xii
11:10 a.m. Report on the U.S. -Japan Cooperative Research Program on Large-Scale

Testing —

11:35 a.m. Report on the U.S. -Japan Cooperative Research Program in Earthquake

Engineering on Repair and Retrofit Buildings and Lifelines —

Task Committee Reports

1: 30 p .m. A. Strong-Motion Instrumentation Arrays and Data

1: 40 p .m

B. Large-Scale Testing Programs

1: 50 p .m. C. Repair and Retrofit of Existing Structures

2: 00 p.m. D. Evaluation of Performance of Structures

2: 10 p .m. E. Land Use Programs for Controlling Natural Hazard

2: 30 p.m. F. Disaster Prevention Methods for Lifeline Systems

2: 40 p.m. G. Extreme Winds and Structural Response

2: 50 p .m

H. Soil Behavior and Stability During Earthquakes

3: 00 p.m. I. Storm Surge and Tsunami

3: 30 p .m

Adoption of Formal Resolutions

4: 00 p .m. Adjourn

xiii
U.S. PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1980

Dr. Edward 0. Pfrang

Chairman

Chief, Structures and Materials Division

Center for Building Technology, NEL National Bureau of Standards

Washington, DC 20234

(301) 921-2196

Dr. H. S. Lew

Secretariat

Leader, Construction Safety

Structures and Materials Division

Center for Building Technology, NEL

Washington, DC 20234

(301) 921-2647

Dr. S. T. Algermissen

Office of Earthquake Studies

Denver Federal Center

Branch of Earthquake Tectonics, USGS

Stop 978, Box 25046

Denver, CO 80225

(303) 234-4014

Dr. Celso Barrientos

Research Meteorologist

National Oceanic and Atmospheric

Administration

8060 13th Street

Silver Spring, MD 20910

(301) 427-7613

Mr. Billy Bohannan

Assistant Director

Wood Engineering Research

Forest Products Laboratory, USDA

P.O. Box 5130

Madison, WI 53705

(608) 257-2211

Dr. Roger D. Borcherdt

Chief, Branch of Ground Motion

and Faulting

Office of Earthquake Studies, USGS

345 Middlefield Road

Menlo Park, CA 94025

(415) 323-8111

Dr. A. Gerald Brady

Physical Scientist

Office of Earthquake Studies, USGS

345 Middlefield Road

Menlo Park, CA 94025

(415) 323-8111

Mr. James D. Cooper

Structures and Applied Mechanics Division

Federal Highway Administration

Office of Research HRS-11

Washington, DC 20590

(202) 557-4315

Mr. Jerry Dodd

Bureau of Reclamation

P.O. Box 25007

Denver Federal Center

Denver, CO 80225

(303) 234-3089

Mr. G. Robert Fuller

Architectural and Engineering Division

Department of Housing and Urban Development

Washington, DC 20411

(202) 755-5924

Dr. Michael P. Gaus

Division of Problem-Focused Research

Applications

National Science Foundation

1800 G Street, NW Washington, DC 20550

(202) 632-5700

Mr. John J. Healy

Chief, Research and Development Office

Department of the Army

DAEN-RDM

Washington, DC 20314

(202) 272-0260

Dr. William B. Joyner

Office of Earthquake Studies

Branch of Ground Motion and Faulting, USGS

345 Middlefield Road

Menlo Park, CA 94025

(415) 323-8111

Mr. John W. Kaufman

Atmospheric Sciences Division

National Aeronautics and Space Administratio

Marshall Space Flight Center, AL 35812

(205) 453-3104

Mr. James Lander

National Oceanic & Atmospheric Administratio

Environmental Data Service

National Geophysical and Solar-Terrestrial

Data Center

Boulder, CO 80302

FTS 323-6474 or (303) 449-1000, ext. 6474

xiv
Dr. E. V. Leyendecker

Leader, Earthquake Hazards Reduction

Group

Structures and Materials Division

Center for Building Technology, NRL National Bureau of Standards

Washington, DC 20234

Dr. Shih C. Liu

Program Manager

Division of Problem-Focused Research

Applications

National Science Foundation

Washington, DC 20550

(202) 632-5700

Dr. Richard D. McConnell

Office of Construction

Veterans Administration

811 Vermont Avenue, NW Washington, DC 20420

(202) 389-3103

Dr. William F. Marcuson, III Research Civil Engineer, WES-SH

Waterways Experiment Station,

Corps of Engineers

P.O. Box 631

Vicksburg, MS 39180

FTS 542-2234 or 636-3111, ext. 2234

Dr. Richard D. Marshall

Leader, Structural Engineering

Structures and Materials Division

Center for Building Technology, NEL National Bureau of Standards

Washington, DC 20234

(301) 921-3471

Mr. John F. Meehan

Department of General Services

Office of Architectural Construction

Sacramento, CA 95805

(916) 445-8730

Mr. Howard L. Metcalf

Deputy Director for Construction

Standards and Design

Office of the Assistant Secretary

of Defense

(MRA&L) Room 3E763

The Pentagon

Washington, DC 20301

(202) 695-2713

Dr. John B. Scalzi

Program Manager, Division of ProblemFocused Research Applications

National Science Foundation

1800 G Street, NW Washington, DC 20550

(202) 632-0648

Mr. Charles Scheffey

Director, Office of Research

Federal Highway Administration

Department of Transportation

Washington, DC 20590

(202) 357-5754

Mr. Lawrence C. Shao

Assistant Director for General Reactor

Safety Research

Nuclear Regulatory Commission

Washington, DC 20555

(301) 427-4442

Dr . Warren A. Shaw

Head, Civil Engineering Department

Civil Engineering Laboratory

Naval Construction Battalion Center

Port Huenerae, CA 93043

(805) 982-5407

Dr. Charles C. Thiel

Assistant Associate Director for Mitigation

and Research

Federal Emergency Management Agency

Washington, DC 20472

(202) 653-7860

Alternates

Mr. Michael Changery

National Climatic Center

Federal Building

Asheville, NC 28787

(704) 258-2850, ext. 217

Dr. Jerry Harbour

Site Safety Research Branch

Office of Site Safety Research Branch

Nuclear Regulatory Commission

Washington, DC 20555

(301) 427-4370

Dr. R. B. Matthiesen

Research Civil Engineer

Branch of Seismic Engineering, USGS

345 Middlefield Road

Menlo Park, CA 94025

Mr. Ronald J. Morony

Program Manager

Building Technology Research Staff

Department of Housing and Urban

Development

Washington, DC 20410

(202) 755-0640

Mr. Drew A. Tiedemann

Bureau of Reclamation

Engineering and Research Center

Denver Federal Center

Denver, CO 80225

(303) 234-3029

xv
JAPANESE PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1980

Mr. Yoshiziro Sakagami

Chairman

Director General, Public Works

Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi

Tsukuba-gun

Ibaraki-ken 305, Tokyo

Dr. Ryuichi Iida

Secretariat

Director, Planning and Research

Administration Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi,

Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Tatsuo Asama

Director, Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi,

Tsukuba-gun

Ibaraki-ken 305, Tokyo

Dr. Nozorau Den

Head, Seismology and Volcanology Division

Meteorological Research Institute (MRI)

Japan Meteorological Agency (JMA)

1-1, Nagamine, Yatabe-machi , Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Shigemi Fjuiwhara

Head, Typhoon Research Division

Meteorological Research Institute (MRI)

Japan Meteorological Agency (JMA)

1-1, Nagamine, Yatabe-machi, Tsukuba-gun

Ibaraki-ken 305, Tokyo

Dr. Yoshimi Goda

Director, Marine Hydrodynamics Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagese, Yokosuka-shi

Kanagawa-ken 239, Tokyo

Mr. Hiroshi Hashimoto

Head, Coastal Engineering Division

River Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi,

Tsukuba-gun

Ibaraki-ken 305, Tokyo

Dr. Sadaiku Hattori

Chief, Seismology Section

International Institute of Seismology and

Earthquake Engineering (I.I.S.E.E.)

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi , Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Masaya Hirosawa

Chief Researcher

Building Research Institute

Ministry of Construction

Tagehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Toshio Iwasaki

Head, Ground Vibration Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Motoo Kobayashi

Head, The First Geographic Division

Geographical Survey Institute

Ministry of Construction

Kitazato 1-banchi, Yatabe-machi, Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Eiichi Kuribayashi

Director, Earthquake Disaster

Prevention Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Tatsuro Murota

Chief, Building Aerodynamics Section

Structure Engineering Division

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305, Tokyo

xvi
Dr. Keikichi Naito

Head, Meteorological Satellite Division

Meteorological Research Institute (MRI)

Japan Meteorological Agency (JMA)

1-1, Nagamine, Yatabe-raachi , Tsukuba-gun

Ibaraki-ken 305, Tokyo

Dr. Kiyoshi Nakano

Director, Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi

Tsukuba-gun

Ibaraki-ken 305, Tokyo

Dr. Nobuyuki Narita

Head, Structure Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi

Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Setsuo Noda

Chief, Earthquake Disaster Prevention

Laboratory

Structure Division

Port and Harbour Research Institute

Ministry of Transport

3-1-3 Nagase, Yokosuka-shi

Kanagawa-ken 239

Dr. Tadayoshi Okubo

Deputy Director-General

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi,

Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Keiichi Ohtani

Chief, Earthquake Engineering Laboratory

Second Research Division

National Research Center for Disaster

Prevention

Science and Technology Agency

3-1, Tennodai, Sakura-mura, Niihari-gun,

Ibaraki-ken 305, Tokyo

Mr. Michio Ohtsuka

Acting Director, International Institute

of Seismology and Earthquake Engineering

(I.I.S.E.E.)

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi,

Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Yasushi Sasaki

Head, Soil Dynamics Division

Construction Method and Equipment Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Yukitake Shioi

Head, Foundation Engineering Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305, Tokyo

Mr. Hiroshi Takahashi

Head, Second Research Division

National Research Center for Disaster

Prevention

Science and Technology Agency

3-1, Tennodai, Sakura-mura, Niihari-gun

Ibaraki-ken 305, Tokyo

Mr. Hajirae Tsuchida

Chief, Earthquake Resistant Structures

Laboratory

Structure Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi,

Kanagawa-ken 239, Tokyo

Mr. Osamu Ueda

Head, Earthquake Engineering Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305, Tokyo

Dr. Makoto Watabe

Head, Structure Division

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305, Tokyo

xvii
RESOLUTIONS OF THE TWELFTH JOINT MEETING

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS

U.J.N.R.

May 19 - 22, 1980

Washington, DC U.S.A.

The following resolutions for future activities of this Panel are hereby proposed

(adopted )

1. The Twelfth Joint Meeting was an extremely valuable exchange of technical information

which was beneficial to both countries. In view of the importance of the cooperative

programs on the subject of wind and seismic effects, the continuation of Joint Panel

Meetings is considered essential.

2. The exchange of technical information, research data, the promotion of joint research

programs including the exchange of personnel, and the use of available facilities

should be strengthened.

3. In order to facilitate the activities of the Task Committees, the nomination of Task

Committee vice-chairmen and expansion of the task committee membership is encouraged.

To expand the activities of the Task Committee G, it is proposed that the name of the

Task Committee be revised to "Wind Characteristics and Structural Response," and to include additional members.

4. The Panel on Wind and Seismic Effects recognizes the accomplishments of the U.S.-Japa

Cooperative Program on Large Scale Testing. This Panel will consider at an early

date, possible endorsement of the Program upon review of the proposals of both sides.

Close working relationships should be maintained between the Panel and the Joint

Coordinating Group of the Large Scale Testing Program. This should include notification of proposed research and change in direction and a periodic review of the

progress of the Program.

5. The Panel on Wind and Seismic Effects recognizes the importance of the U.S. -Japan

Cooperative Program on Repair and Retrofit of Buildings and Lifeline. It recommends

expended implementation of the program through this Panel. The Panel urges that the

program coordination of both sides shall maintain close working relationships with

this Panel and the Secretariats of both sides shall be kept informed of all

correspondence

6. Recognizing that wind and seismic disaster mitigation plans benefit disaster-prone

countries worldwide, that many developing countries suffer from natural disasters

such as extreme winds and earthquakes, and that both the United States and Japan

have been involved in international technical cooperation programs, both sides of

this Panel should give consideration to conduct coordinated projects for the aid of

developing countries and to maintain exchange of available technical information.

7. The date and location of the Thirteenth Joint Meeting of the Panel on Wind and

Seismic Effects will be May 1981 at Tsukuba Science City, Japan. Specific dates

and itinerary will be proposed by the Japanese Panel with concurrence by the U.S. Panel

xviii
DENSE INSTRUMENT ARRAY PROGRAM OF THE PUBLIC WORKS

RESEARCH INSTITUTE FOR OBSERVING STRONG EARTHQUAKE-MOTION

by

Yoshiziro Sakagami

Tadayoshi Okubo

Toshio Iwasaki

Kazuhiko Kawashima

ABSTRACT

Based upon the knowledge accumulated from past strong-motion observations, it is

widely recognized that short period ground motions are significantly dependent on source

characteristics, path conditions between the source and the observation station, local geological and topological conditions. In order to investigate these effects, normal strongmotion observations performed at individual sites are not enough, and installation of

dense instrument arrays are considered necessary.

Solving these important earthquake engineering problems, the Public Works Research

Institute intends to press for research funds to further the dense instrument array program

at ten locations. To prepare for these observations, the Public Works Research Institute

installed a simple extended array consisting of four strong-motion accelerographs as a

pilot station at Ashitaka Area, Shizuoka Prefecture, to begin the investigation of local

soil condition effects on strong ground motions. A local laboratory array installed at

the Public Works Research Institute is also being employed for this purpose.

This report briefly summarizes the dense Instrument array program of the Public Works

Research Institute and details the preliminary pilot array observations at PWRI in Tsukuba

Science City and at Ashitaka Area, Shizuoka Prefecture.

i
DENSE INSTRUMENT ARRAY OBSERVATIONS — BACKGROUND

In the earthquake engineering field, strong motion observations have been made in an

effort to collect the actual data of seismic responses of structures and subsurface ground

nearby so that improved seismic design procedures for various kinds of structures can be

achieved from analysis of the recorded responses. Observations have mainly concentrated

on acceleration measurements. In Japan strong-motion observation began around 1955. They

were activated following the Niigata-Earthquake in 1964 in recognition of the importance of

strong-motion records for analyzing earthquake damage. Currently more than a thousand

strong-motion accelerographs have been installed which cover the whole of Japan.

As far back as 1934 observations began in Japan using elementary arrays of several

instruments extending unilaterally in the vertical direction. Practical observations using

elemental arrays began after 1959, and now the number of observatory points has reached

about one hundred. The purpose of these observations has been the analysis of structural

response and soil-structure interaction, while some of them have been installed for the

analysis of source mechanism, wave propagation, and local site effects.

Based on the knowledge accumulated from past strong motion observations, it is widely

recognized that the characteristics of ground motions, especially ground motions of short

period, are strongly dependent on source characteristics, path conditions between the

source and the observation station, local geological and topological conditions. It is

also recognized that in order to investigate these effects, normal strong motion observations performed widely at the present are not enough, and that the installation of dense

arrays are necessary.

Based on this historical background, the Executive Committee of the International

Association for Earthquake Engineering, meeting on January 14, 1977 in India on the occasion of the Sixth World Conference on Earthquake Engineering, decided to hold an International Workshop on Strong-Motion Earthquake Instrument Arrays under the sponsorship of

UNESCO. The International Association of Seismology and Physics of the Earth's Interior

(IASPEI) joined this effort as a member of International Workshop on Strong-Motion

Earthquake Instrument Arrays.

The International Workshop was held at Honolulu, Hawaii in May 1978 with participants

from sixteen countries. In Japan, before holding of the workshop, another workshop on

2
Strong-Motion Earthquake Instrument Arrays was organized by the Steering Committee of

Earthquake Engineering, Japan Science Council, to prepare a Japanese master plan of array

observation. This Japanese workshop has continued to serve in the forming of Japanese

plans and coordinating proposals from participating organizations.

The international workshop recommended the formation of an International Strong-Motion

Arrays Council to facilitate the establishment of strong-motion earthquake instrument

arrays. Based on a careful examination of the potential for the occurence of large earthquakes in 10 years, 28 locations were selected as favorable array locations. Of the 28

locations selected, six have been designated as high priority sites. These sites were

judged to have an especially high probability of yielding useful data. Among the 28 locations, Eastern Tohoku, Suruga Bay-Izu, and Western Chubu areas were included, and the

Suruga Bay-Izu area was designated as one of the six high priority sites.

The international workshop also recommended typical array configurations based on the

fault model, such as strike slip fault, subduction thrust fault, and dip slip fault for

analysis of source mechanism and wave propagation, and on topological and geological conditions for analysis of local site conditions. The establishment and maintenance of mobil

strong-motion instrument array capable of making source mechanism wave propagation, and

local effects measurements was also recommended for deployment immediately after the occurence of major earthquakes to record aftershocks.

Following the recommendations of the international workshop, a program of deployment

of strong-motion earthquake instrument arrays in Japan is now being developed by the

Japanese workshop. The program recommends three items, i.e., (a) development of a strongmotion earthquake instrument arrays system, (b) installation of arrays for source mechanism

and path effect studies, and (c) installation of arrays for local site effects studies. It

is hoped the program will be realized by both scientific and engineering groups. Participants include the: School of Engineering, University of Hokkaido, Earthquake Research

Institute, University of Tokyo, School of Science and Engineering, The University of Tohoku

and The University of Nagoya, Disaster Prevention Research Institute of the University of

Kyoto, Public Works Experimental Station of the Hokkaido Development Bureau, Meteorological

Research Institute of the Meterological Agency, National Research Center for Disaster

Prevention of Science and Technology Agency, Public Works Research Institute of Ministry

3
of Construction, Building Research Institute of Ministry of Construction, and Port and

Harbour Research Institute of Ministry of Transportation. Also, it is expected that the

Japan Science Council will recommend that the Prime Minister of the Japanese Government

enthusiastically promote the Implementation of dense strong-motion instrumental array

programs, including funding support.

THE DENSE INSTRUMENT ARRAY PROGRAM OF THE PUBLIC WORKS RESEARCH INSTITUTE

Recognizing the importance to understanding the characteristics of strong-motions In

order to improve seismic design procedure, the Public Works Research Institute has conducted strong-motion observations since 1957 for bridges, dams, river dykes, estuary dams,

underground shopping streets, submerged tunnels, et al., and the surrounding subsurface

ground. Up to March 1980, the total number of SMAC accelerographs and electro-magnetic

accelerographs reached 261 (131 observation stations) and 46 (46 observation stations,

439 channels) , respectively. Measurements of underground motions also have been conducted

since 1970 using down-hole accelerographs installed as deep as 120 m below the ground surface at seven locations in the bay areas, of Tokyo, Osaka, and Nagoya. A local laboratory

array deploying accelerographs in horizontal and vertical directions is being equipped at

the Public works Research Institute in Tsukuba Science City and is expected to begin

measurement after May 1980.

A fuller understanding of the nature of strong ground motions is needed to provide

answers to unsolved problems in estimating rational input ground motions to large structures

such as long suspension bridges and submerged tunnels, in evaluating seismic risk to lifeline network facilities, and in developing suitable seismic micro-zoning procedures applicable to earthquake disaster prevention planning.

Considering the importance of solving these problems, the Public Works Research

Institute intends to press for research funds to promote dense instrument array programs

at the ten locations shown in Figure 1, which will deploy at least ten strong-motion accelerographs at each location. Figure 2 shows a proposed standard array configuration. To

prepare for these practical observations the Public Work Research Institute installed a

simple extended array of four strong motion accelerographs at the pilot station in the

Ashitaka Area, Shizuoka Prefecture. Observations to begin about May 1980.

4
THE LOCAL LABORATORY ARRAY AT THE PUBLIC WORKS

RESEARCH INSTITUTE IN TSUKUBA SCIENCE CITY

Location

The Public Works Research Institute in Tsukuba Science City is located approximately

60 km northeast of Tokyo as shown in Figure 3. Figure 4 shows the research facilities of

the institute. Two local laboratory arrays are located at areas A and B, which are designated hereafter as A Field and B Field. At C area (designated as C Field) a strong-motion

accelerograph with an autodial data transmission system is installed.

Geological conditions around the institute are almost uniform in the horizontal

direction as shown in Figure 5. Alluvial sandy and silty deposits approximate thickness

of 50 m, rest on a diluvial gravel formation as shown in Figure 6. Shear wave velocities

of the alluvial and diluvial are approximately 250 m/ sec and 400 m/ sec , respectively.

Instrumentation

Figure 7 shows instrumentations at the A, B, and C Fields. In the A Field are 13

three-component accelerometers , i.e., seven on the surface, one at a depth of 2 m and 5

at a depth of 50 m, along an x-shaped configuration, with leg length 100 m. In the B-Field

are 6 three-component accelerometers, i.e., one at the depth of 2 m, four at the depth of

66 m and one at the depth of 96 m, along an L-shaped configuration of short and long

lengths of 50 m and 100 m, respectively (refer to Table 1). The x and L-shaped configurations are orientated along NS and EW lines, as are the sensor components. The main specifications of the accelerometers used are summarized in Table 2. Signals from each sensor

are transmitted by cable placed in a duct as shown in Figure 8 to a central processing

room of strong-motion earthquake records via a field station house. Figure 9 shows an

aerial view of the A and B Fields.

At the C Field, an accelerometer is set on the ground surface. A self-trigger system

is provided. Specifications of the accelerometer are essentially the same as those used

in the A and B Fields.

At Fields A and B, signals are triggered either when maximum acceleration of vertical

component exceeds 3 gals at both Fields (one of points with depth of 50 m at A Field, and

5
a point with depth of 96 m at B Field) or when maximum acceleration of one of two horizontal components at both points exceeds 5 gals. At C Field triggering of the records occurs

when maximum acceleration of vertical component exceeds 3 gals.

Recording and Processing Systems

Signals from the sensors installed at A, B, and C Fields are recorded on a computer

system magnetic tape in the following form (refer to Figure 10):

(i) Excluding 21 signals of the surface accelerometers at A Field, 36 of the total

57 signals triggered at A and B Fields are automatically digitized with a time

interval of 1/100 second by an AD converter and stored on magnetic disk (maximum capacity of 64 MB) by the 80 KW core memory. Pre-event memories of 5

seconds are prepared for the 36 signals so that an entire recording of earthquakes motions can be achieved. Data stored on the magnetic disk are then

compiled into a form of standard data processing of strong-motion records

formulated by the Public Works Research Institute, and stored on a magnetic

tape. Besides this recording system, an analog tape recording system is

supplementarily provided to store 32 of the the 36 signals. This system can

be used when the computer recording system is out of use.

(ii) While the 21 signals of the surface accelerometers at A-Field are also

automatically digitized with a 1/100 second time interval by AD converter and

stored on magnetic disk, pre-event memories are not provided. But time code

simultaneity of the 21 signal group with those of the 36 signals is kept,

(iii) The accelerometer signal at C-Field is automatically digitized with a 1/100

second time interval by AD converter and stored on a digital cassette tape. A

prevent memory of 10 seconds is provided. After an earthquake, the transmitter

at C-Field automatically seeks, via telephone cable, an extension of the central

processing room up to three times. If a connection occurs within any one

of the three calls, the recorded data are transmitted to the computer system

magnetic disk in the central processing room. If no connection is achieved

within three calls from the transmitter, the transmitter goes into a hidden

6
stage of keeping the data stored on the cassette tape. In such a waiting

situation, the transmission of the data from cassette tape to magnetic disk

can be achieved by calling back from the central processing room to the

transmitter at C-Field. A backup cassettee tape is provided to work so that

signals induced by the next earthquake can be recorded even in cases when the

first cassette tape is full and on hold.

In the recording system above, the computer system plays a principal role,

and when the system is down, no recording of data can occur. Therefore, a

float battery charger will supply system power for no less than 20 minutes if

the electrical lines are knocked out.

For analyzing the recorded motions, a 192 KW core memory computer system

is provided as shown in Figure 11. Data transmission from recording system

to analyzing system is performed on a tape base. The system is designed to

analyze not only the data from the A, B, and C Fields in Tsukuba Science City,

but also the data obtained through the strong-motion observation network of

the Public Works Research Institute which covers Japan.

THE SIMPLE EXTENDED ARRAY AT ASHITAKA AREA IN SHIZUOKA PREFECTURE

Location and Instrumentation

The Ashitaka Area is located between Numazu City and Fuji City in Shizuoka Prefecture

facing Suruga-Bay. Figures 12 and 13 show the geological surface conditions and a cross

section of the geological profile in the NS direction. The surface of the Ukishimagahara

,

Ashitaka Area, is covered with organic soil as deep as 15 m, and is underlaid by gravels,

sandy and silty layers. The baserock of this area is tuff breccia existing approximately

at a depth of 160 m below the surface at Higashi-tago-no-ura

.

Figure 14 represents one of the typical soil profiles at Ukishima-gahara in the Ashitaka

Area. Shear wave velocity of the subsurface material is in the range of 70 to 200 m/sec.

Tuff breccia exists at a depth of approximately 35 m below the surface and has a shear

wave velocity of approximately 650 m/sec at this site.

7
A simple extended array is being installed by the Public Works Research Institute

crossing Ukishima-gahara in an almost NS direction as shown in Figure 15, which also presents a geological section obtained from boring data. Japan's most important traffic facilities, the National Highway Route 1, bi-path of Route 1, Tokaido Line (Railway) and

Shinkansen (New Tokaido Railway), cross the array in an EW direction.

The array consists of five strong-motion accelerographs ; four are installed already,

i.e., one on a sand hill facing Suruga Bay, two on Ukishima-gahara, and one on a foot of

the Ashitaka Mountain. An additional accelerograph is being installed between the

Ukishima-gahara and a foot of Ashitaka Mountain. Figure 16 shows an aerial view of the

array location.

On-site Accelerograph Storage Buildings

Four RC on-site buildings, as shown in Figure 17, were constructed to store the

accelerographs at each location. Each accelerograph is set on a concrete footing 1.2 m

in length, 12 m in width, and 0.3 m in thickness, and is isolated from the building

footing as shown in Figure 18. Nine pine piles 20 cm in diameter and approximately 2 m

long are attached to the bottom of the footing which supports the accelerographs to

avoid the development of relative movement between the footing and the surrounding

ground during earthquakes. The vertical position of the accelerographs is designed

to be 35 cm higher than the surface of the concrete footing to avoid inundation of the

accelergraphs in case of flooding.

Recording System

Four 3-component strong-motion accelerographs using digital cassette recording systems

are employed for the array. Accelerograph specifications are summarized in Table 3. A maximum acceleration of ± 1 g can be recorded, the signals are digitized at 1/200 second time

intervals, and are cassette stored. Time code generators are equipped for each recording

system so that exact triggering time is recorded at the four sites. Crystal accuracy in

the time code generator is approximately ± 10"~6 (3 seconds per month). Auto adjustment

of the crystal can be performed each 5 seconds using JJY signal. Time code generator error

is designed to be controlled less than 10~3 second.

8
Float battery chargers supply electricity for no less than 6 hours in case of

electricity breakdown.

AN EXAMPLE OF LOCAL LABORATORY ARRAY RECORDING

Both the local laboratory array at PWRI in Tsukuba Science City and the simple

extended array at Ashitaka Area are scheduled to begin full observation after May 1980.

Therefore no records have as yet been obtained at either array. A preliminary and limited

observation has been performed since July 1979 on the local laboratory array using an

auxiliary analog tape recording system, and a record induced by an earthquake of July 11,

1979, Richter Magnitude of 5.9, focal depth of 40 km, and an epicentral distance of

approximately 120 km was successfully recorded.

Figure 19 represents acceleration records thus obtained at the north, east, south, and

west edges of the cross at the 50 m depth. These are designated as N50, E50, S50, and W50

(refer to Figure 20). Since the temporary recording system consists of analog tape, simultaneity of time code is achieved between the records at N50 and W50, and between the

records at E50 and S50 only.

Figures 21 and 22 represent power spectra of the acceleration records and

displacements computed by double integration of the accelerations, respectively.

Ground strains induced by the earthquake were estimated using the displacements shown

in Figure 22. An averaged ground strain between two observing points is defined as

e Ax

eSH

10 -10

0 1 0-1

"2,1

(1)

where Eaw and egjj represent averaged axial strain and shearing strain between the specified

two points, respectively, and where \i\ ^ and \i2 k (k=l, j) represent ground displacement at

point k in NE-SW and NW-SE directions, respectively. A parameter L is defined by a length

between the two points and is given by /2 I in this case, where I equals 50 m.
The displacement and P2,k (^=1, j) can be determined in the foi

W1,J

U2,i

U2,j

1

_ 1 1 0 0- , ^NS ,

i

-1 1 0 0 1 ^EW.i

0 0 1 1 )TMS, j

- 0 0--1 L-

(2)

where MNS.k anc* ^EW k (k=i,j) represent displacements at point k in the N-S and E-W

directions, respectively.

According to Eq. (1), the averaged axial strain e^x an<* shearing strain eg^ were

computed between N50 and W50, and between E50 and S50 as shown in Figure 23. The maximum

value of both and e^^ are estimated as 5 x 10

-

^. One can approximate the averaged

strain from displacement represented in Figure 22 assuming that the displacements on

points i and j are in out of phase. Because the maximum displacements in Figure 22 are

approximately 0.3 cm, the averaged strain based on the above assumption can be estimated as

2 ^max 0.3 i.5 x 10-5 L ST x 5 . 000

which is close to the strains determined by Eq. (1).

(3)

CONCLUDING REMARKS

The dense instrument array program of the Public Work Research Institute has been briefly

described with emphasis on preliminary pilot array observations conducted at PWRI in

Tsukuba Science City and at Ashitaka Area in Shizuoka Prefecture. The dense instrument

array program has just begun in Japan, and because of the importance of investigating local

effects on strong-ground motion in the earthquake engineering field, the encouragement of

this program is highly recommended.

REFERENCES

[1] International Association for Earthquake Engineering: Strong-Motion Earthquake

Instrument Arrays, Proceedings of the International Workshop on Strong-Motion Earthquake Instruments Arrays, Edited by W. D. Iwan, 1978, Honolulu, Hawaii, U.S.A.

[2] Iwasaki, T. , Wakabayashi, S. and Tatsuoka, F. : Characteristics of Underground

Seismic Motions at Four Sites Around Tokyo Bay, 8th UJNR, Washington, D.C., 1976.

10
Nakano , S. and Kitagawa, Y. : Earthquake Observation Systems In and Around Structures

in Japan, 11th UJNR, Tokyo, 1979.

Shizuoka Prefecture: Geology in Shizuoka Prefecture, edited by Tsuchi, R. , 1974.

Development Bureau of the Economic Planning Agency: Land Classification Map, No. 08,

Part of Ibaragi-Prefecture , 1973.

Kuribayashi, E., Tsuchida, H. and Watabe, M. : On the Maintenance of the StrongMotion Accelerograph and the Data Processing the Records, 7th UJNR, Tokyo, 1975.

Okubo, T. : Dense Instrument Array Program, Civil Engineering Journal, Vol. 22, No. 4, 1980.

1 1
Fig. 1 Proposed Location of Dense Strong-Motion Array by PWRI

12

14
SAND i lAM GABBRO

SERIES GNEISS .EGEND

MUD

I

GRANITE BLACK SCHIST CLASTICS

Fig. 5 Surface Geological Map Around Public Works Reserach Institute

at Tsukuba Science City (After Reference 5)

1 5
0
B FIELD

Fig. 7 Local Laboratory Arrays at PWRI

17
Table 1 Number of Accelerometer and Number of Channel

of Strong-Motion Arrays at PWRI

FIELD

1 DEPTH POINTS CHANNELS

GROUND

7 21 SURFACE

A-F1E1.D

1 -2m 1 3

-50m 5 15

-2m 1 3

P-FIEI D -66m

1 U 12

-96m > 3

C-FIELD GROUND

SURFACE

1

1 3

TOTAL 20 60

Table 2 Main Specification of Accelerometer used

by Local Laboratory Array at PWRI

Number of Component 3

Type Velocity Feed Back

Natural Frequency 5 Hz

Frequency Range 0.1 - 50 Hz ; within 1 dB

Maximum Acceleration + 650 gals

Diameter, Length i 13 cm, 125 cm

Weight 50 kg

Fig. 8 A-Field of Local Laboratory Array

in PWRI at Tsukuba Science City

i 8
19
36 ACCELEROMETERS O-r-0

21 ACCELEROMETERS 1 ACCELEROMETER

DATA PROCESSING ROOM

A/D CONVERTER

DEIAY_ MEMORY

b seconds"

(MAX 4JCHS i

Or©

|_ A/C

PIC

CONVERTER ! PROCESSOR

i

TWO CASSETTE TAPES.

_ CPU

IC MEMORY H'iKvv MAGNETIC DISK 6^MB 3 ©

Fig. 10 Data Processing System

CARD [DIGITIZED STRONG

[MOTION ACCERELATION

CPU

CORE MOMORY

192Kw MAGNETIC DISK 10MB

POWER SPECTRA. FOURIER SPECTRA, RESPONSE SPECTRA. FREQUENCY

I RESPONSE FUNCTION, BASE LINE

' CORRFCTION. INTEGRATION

Fig. 11 Data Analyzing System

20
m

EM

mm

+ -n

LEGEND

GRAVAL

SAND

MUD

SAND, GRAVEL AND MUD

LOW TERRACE DEPOSIT

HIGHER TERRACE DEPOSIT

EARLY DILUVIAL DEPOSIT

VOLCANIC ASHfSHIRAHAMA GROUP

etal) VOLCANIC ASH(YUGASHIMA GROUP

etal) ANDESITE

VOLCANIC ASH AND LOAM

LAVA

LAVA

EFFUSIVE ROCK FROM OLD FUJI BASALTIC EFFUSIVE ROCK

PUMICE FLOW

OLD LAVA OF SOMMA

BASALTIC LAVA

ANDESITE LAVA(DARUMA VOLCANO)

ANDESITE LAVAflDA VOLCANO)

DECITE

Fig. 12 Surface Geological Map Around Ukishima-Grara

,

Ashitaka Area, Shizuoka Prefecture

21
HIGASHI TAGO NO URA

' B ORGANIC SILT OR CLAY

: SAND

i FZD GRAVELS

~ LUI VOLCANIC GRAVELS

5 m I OAM

"S3 TUFf BRECCIA

Fig. 13 Geological Condition Around Ukishima-garaha , Ashitaka Area

(After Geology in Shizuoka Prefecture edited by Tsuchi)

22
(7) Ashitaka Junior High School

+20m

Railway

(Tokai-do Shmkansen)

+ 10m

Bong No

3302

(T\ Imazawa (^National Hiqhway

\2/ ci «t^„, c^v.^^i Eienientary School (Bi - Path ol Routo

Railway National Highway

(Tokaido Line) W ( R OU | e i,

Bong No

3075

Bong No BonqNo Bonq No Bong No

3026 3027 3016 3024

-30m

LEGEND

^<T1 TOP SOIL

[gg] GRAVELS

[~1 SAND

SILT OR CLAY

ORGANIC SOILS

SILTY SOILS

fTJ^H LOAMY SOILS

Fig. 15 Geological Condition for Simple Extended Array at Ukishima-gahara , Ashitaka Area

23
Fig. 17 On-Site Housing Storing Accelerograph

Fig. 18 On-Site Housing Storing Accelerograph

24
(b) E50 NS-Component

EW-Component

UD-Component

(d) W50

Fig. 19 Strong-Motion Accelerations Recorded at N50, S50, E50 and W50 of I.oca Laboratory Array of PWRI at Tsukuba Science City
Table 3 Main Specifications of Strong-Motion Accelerograph Used for

Simple Extended Array at Ukishima-gahara , Ashitaka Area

pNMI

AXIAL STRAIN BETWEEN E50 AND S50 : eAX SHEAR STRAIN BETWEEN E50 AND S50: e s „

J 2 1/ 10-10

0 1 0-1

y, £su

M y,6'50

Accelerometer

Type Triaxial, Force Balance

FuLl Scale Range t 1C

Natural Frequency 50 Hz

Damping 70% critical

Trigger System

Sensitive Direction Vert ic3l

Acceleration Set Point 0.01C

Operation Cycle Self-actuating for duration of earthquake, with automatic reset

Recording System

Type Digital, phase encoded

Number of Truck 4

Mangetic Tape Digital cassette tape; 2.5inch/

sec

Frequency Response DC - 50 Hz; -12 dB/oct

Number of Bits 12 bits

Dynamic Range f 66 dB

Sampling Rate 200 Samples/sec/channel

Power Requirements

Voltage + 12 VDC

Battery Charger Float charger supplied

Time Code Generator

Accuracy of Crystal - Controlled Time ± 1 x 10" 6

Auto Adjustment of Time of Crystal

Each 5 seconds by JJY Standard

Time

cos# sin# 0 -sm0 cos# 0 0 0 1

Fig. 20 Definition of Axial and Shear Strains Between

Two Points

26
- 'E50

- > S50

- - W50

- • N50

0 S . C i 0.

0

FREQUENCY !HZ) (a) NS-Component

-E50

-S50

-W50

-N50

FREQUENCY

(b) EW-Coraponent

b. 0 r E5 0 •S50

'W50

0. 010 0 1.0 2.0 3.0 4.0 5.0 &.0 7.0 8.0 9.0 10.0 FREQUENCY (HZ)

(c) UD-Component

Fig. 21 Power Spectra of Acceleration Records Obtained at N50, S50, E50 and W50

27
- V /"v. NS-Component

EW-Component

(a) N50 NS-Component

(b) E50

(c) S50

NS-Component

NS-Component

EW-Component

(d) W50

Fig. 22 Displacement Time History Calculated by Double Integration of Strong-Motion

Acceleration Records at N50, E50, S50 and W50 of Local Laboratory Array of PWRI at Tsukuba Science City

(b) Shear Strain Between N50 and W50

(c) Axial Strain Between E50 and S50

(d) Shear Strain Between E50 and S50

Fig. 23 Axial and Shear Strains Between Two Points

28
U.S. GEOLOGICAL SURVEY STRONG-MOTION RECORDS FROM THE IMPERIAL

VALLEY EARTHQUAKE, OCTOBER 15, 1979, PRELIMINARY SUMMARY

by

J. R. Filson

R. L. Porcella

R. B. Matthiesen (Deceased)

This report summarizes the data from near-in strong-motion accelerograph stations

operated by the U.S. Geological Survey (USGS) in the Imperial Valley of California at the

time of the October 15, 1979 Imperial Valley earthquake. The purpose of this report is to

alert others as to the nature of the strong-motion data that is available from this event.

In order to provide the information in a timely manner, the report has been limited to a

summary of the data. A similar preliminary report of strong-motion data collected by the

Office of Strong-Motion Studies of the California Division of Mines and Geology (CDMG) has

already been issued.* A more complete report of all of the strong ground motion data is

contemplated. This will require the cooperation of all of the agencies in both the U.S.

and Mexico that operate strong-motion instruments in the region. A report on the

processing of the data from near-in stations is in preparation.

The moderate-size (magnitude 6.4) October 15, 1979 Imperial Valley earthquake was

instrumentally located on the Imperial fault approximately 25 km southeast of El Centro,

California (figure 1). This location is approximately the same as that of the 1940

Imperial Valley earthquake. The following parameters are for the main shock (CIT/USGS):

Origin time: 23:16:52.4 15 OCT 79 (UTC)

Epicenter: 32.64 N, 115.33 W

Focal depth: 15 km

Magnitude: 6.4 (ML )

The 1979 Imperial Valley earthquake triggered all of the accelerographs within about

100 km of the epicenter, and one as far away as 196 km. At this time (28 OCT 79) information is not available on all of the stations beyond 150 km. The locations of the ground

motion instruments relative to the epicenter are shown in figure 1. A list of all of the

* Preliminary Data - Partial film records and file data - Imperial Valley Earthquake of

15 October 1979; CDMG; Sacramento, 1979.

29
accelerographs that are known to have been operational at the time of the earthquake and

within a radius of about 150 km of the epicenter is given in table A. This list contains

the station identification, an indication of the site geology, and the type and size of

structure in which the instrument is housed. Most of the instruments are located in small

instrument shelters or one-story buildings and are assumed to approximate a "free-field"

condition. The instruments are self-contained film-recording accelerographs with a nominal

upper limit on range of 1.0 g.

The data from USGS operated accelerographs are summarized in Table B and are

presented in order of increasing distance of the stations from the 1940 fault trace. In

addition to the distance from the fault and epicenter, the table summarizes the S-wave

minus trigger times (when discernable) , the peak accelerations, strong durations, and the

trigger time as indicated by the WWVB time code on the record. Copies of all main-shock

accelerograms from the USGS network are included in Appendix I of the complete report, but

are not included in this summary version.

ACKNOWLEDGMENTS

The USGS acknowledges the cooperation of individuals and organizations that have

permitted strong-motion accelerographs to be installed on their property. The authors

also extend their special appreciation to John Nielson, Ed Etheredge, Leroy Foote,

Dennis Johnson, and Arnie Acosta for their efforts in recovering the records and to

Chuck Knudson and Barry Silverstein for their assistance in verifying the information

about the stations and in scaling the records.

30
Table B - GROUND MOTION DATA

Station Identif ication1 Epicentral^

Distance

S - t3 Interval

( sec

)

WWVB4 triggerAcceleration Duration^

> 0.1 g

ViO • Name

(Data Source) Coord

Azimuth

v note j ) Maximum

5028 El Centro Sta 7 3? 83 w 26 4.6 ** 230 0. 52 4.9

Imperial Valley Col IIS Sn Uw r i l up 0.65 5 .

i An n ifi 3.7

942 El Centro Sta 6 19 ftA M 27 5* 17:01 230 0.45 7.9

Huston Rd 1 1 S AQ TJW r i i up 1.74 ft 9

i An nU • 7?/ Z 1 1 ft

5054 Bonds Corner M 6 2.4 16:57 230 0.81 13.2

Hwys 98 & 115 IIS 34 W r 3

i J J up 0.47 12.0

l ttcpc "\ Ubbb ) 14 u U • DO 11 i

958 El Centro Sta 8

19 01 JZ .51 NT

IN 27 5* 17:00 230 0.50 6.9

95 E Cruickshank 11C COI i J . D i w r A i up 0.55 S ft / TIC PC ^ U bbb )

i An nU • aa04 6.9

952 El Centro Sta 5 MIN 28 5.1 17:01 230 0.40 7.6

2801 James Rd 1 1 S A 7 W r A i

J up 0.71 _> • o

V U bob )

i An U • 30 7 A / • 4

5165 E C Differential Array 32 . 80 N 26 5 ** 360 0.51 10.2

Dogwood Rd 115. 34 W r s i

l -* J up 0.93 7 n 270 0 . 37 i n

117 El Centro Sta 9 3? 79 N 26 — 360 0.40 7.4

302 Commercial Av IIS ss W r 6

L w J dwn 0.38 4.7

090 0.27 7.0

955 El Centro Sta 4

"19 ftA MLN 26 4.8* 17:01 230 0.38 6.5

2905 Anderson Rd 115.43 W r 7 i up 0.32 6.7

f TTC PC "\ ^ Ubbb )

i An n tsi 6.8

5060 Brawley Airport 99 QQ N 42 6.3 17:03 315 0.22 2.2

Brawl ey

IIS ii j si • ji TJw \1 1 L ' J up 0.18 5. 2 (USGS

)

(J . 1 / 1 .

8

5055 Holtville 32 . 81 N 19 4.1 ** 315 0.22 7.5

Post Office 115.38 W r ft i 1° J up 0.31 7 n (USGS) 225 0 . 26 6.2

412 El Centro Sta 10 32.78 N 27 4.9 ** 050 0.20 5.2

Community Hospital 115.57 W r q i l y J up U.n lji s 7 7 (USGS) 320 0.23 C 1 J • 1

5053 Calexico Fire Station 32.67 N 1 s 3 9 1 6 • SQ 315 0.22

Fifth & Mary 115.49 W [11] up 0.21 8.8

(USGS) 225 0.28 10.8

5058 El Centro Sta 11 32.75 N 27 5.6 17:00 230 0.38 6.5

McCabe School 115.59 W [13] up 0.16 7.7 (USGS) 140 0.38 7.0

31
Table B - GROUND MOTION DATA (con't)

Station Identification! Epicentral2

Di s tance

(km)

S - t3 Interval

(sec)

WWVB^

trigger

time

Acceleration Duration"

S n 1 rr > u • 1 g (sec) No. Name

(Data Source) Coord

Azimuth

(note 5)

Maximum

(g)

5057 El Centro Sta 3 32.89 N ** 230 0.22 O . Z

i ILlc U 11 -LU 11 OL. 1 1(J LJ J. 115.38 W [13] up 5.6

(USGS) 140 0.27 6.0

5051 Parachute Test Site 32.93 N LI 7 n ** 315 0.20 i r1 • D

f TTC f~*G "\ { UoOo ) 115.70 W [15] up n 1 q U • 1 o 5.2

225 0.11 1.4

5115 El Centro Sta 2 32.92 N 3 1 o ii 7/

•

. myjL 230 0.43 c; 7 Keystone Rd 115.37 W [16] up U* 1 / 9.3

(USGS) 140 0.33 9.2

931 El Centro Sta 12 32.72 N 3Uin 3 • Z i / : ui 230 0.11 A Q yo/ Brockman Rd 115.64 W [18] up 0. Oo (USGS) 140 0. 15 3.8

5061 Calipatria Fire Sta 33.13 N 3 / 7 /. 17. 315 0.09

(USGS

)

115.52 W [21] up 0. 07 _ 225 0.13 1 peak

5059 El Centro Sta 13 32.71 N

1/, 3. 1 17. no 230 0.15 5 . U Strobel Residence 115.68 W [22] up n esc U • UO (USGS) 140 0.12 2.4

5056 El Centro Sta 1 32.96 N 37 cO

1i 7/

•

• nouz 230 0.15 i i j • i Borchard Ranches 115.32 W [22] up n\J . inIV 1 peak

(USGS) 140 0.15 4.8

286 Superstition Mtn 32.95 N

c:7 3 / 7/ • Z9* 1 7 • (Ts 135 0.21 1 1 1 • 1 UoAr uatQera sice 115.82 W [26] up nU • noU" (USGS) 045 0.12 0.6

5062 Salton Sea 33.18 N DO 31 • 3^*~ 17-11 315 0.10 1 peak

wiiaiire Keruge 115.62 W [28] up u • uo USGS) 225 0.13 1 peak

5052 Plaster City 32.79 N 52 3" ** 135 0.07

Storehouse 115.86 W [31] up U • U3 - (USGS) 045 0.05 -

5066 Coachella Canal Sta 4 33.36 N P.A ftO • ^3 17.11 1 / • 1

1

135 0.14 nU.J^ Siphon 15 115.59 W [47] up 0.04

(USGS) 045 0.11 0.3

5050 Ocotillo Wells 33.14 N 93 7.5* 17:15 315 0.05 _ Burro Bend Cafe 116.13 W [59] up 0.03

(USGS) 225 0.04

2316 Yuma 32.73 N 60 * 17: 11 090 0.03

Strand Ave 114.70 W [61] up 0.02

(USBR/USGS) 360 0.03

32
Table B - GROUND MOTION DATA (con't)

Station Identification! Epicentral2

Distance

(km)

S - t3 Interval

(sec)

WWVB*

trigger

time

Acceleration Duration6

? U . I g (sec) No. Name

(Data Source) Coord

Azimuth Maximum

(note 5) (g)

5065 Coachella Canal Sta 3 33.51 N ins *** jipnon it 115.77 W [67]

(USGS)

5049 Borrego Air Ranch 33.19 N 1 OR O . £a 17-90 315 0. 04

RDUIr-» r" T"LCKLI o rr r\ QnTinoo jpi. XLlgo 116.28 W [74] up 0. 02 - (USGS) 225 0. 03 -

5064 Coachella Canal Sta 2 33.56 N 117 *** Demosser 115.95 W [79]

(USGS)

5047 Rancho de Anza 33.35 N 191Z 7/ 17.97 Li I Z/ 135 0.03

Anza Borrego Park 116.40 W [92] up 0. 02 - (USGS) 045 o 02 -

5063 Coachella Canal Sta 1 33.64 N 1 Jl * 1 7 • 9A 135 0. 02 (USGS) 116.08 W [92] up 0 02 _ 045 0.03 -

5067 Indio 33.75 N 148 ***

So Calif Gas Co 116.21 W [109]

(USGS)

5073 Cabazon 33.92 N 196 * 17:55 270 0 01 Post Office 116.78 W [158] up 0 02 (USGS) 180 0 02

1 Ref: Western Hemisphere Strong Motion Accelerograph Station List - 1976; USGS, Open-File Report 77-374.

2 Distance from epicenter at 32.64 N and 115.33 W. Bracketed number is distance to the nearest point on

the 1940 Imperial Fault trace. Reference: Sharp, R. , 1977; Open File Report 77-815; Misc Field Studies

Map MF 838.

3 S-wave minus trigger time.

* S - t is questionable.

4 Trigger time in minutes and seconds after 288 days (UTC) as determined from WWVB time time code.

Millisecond accuracy is possible.

** WWVB time code not legible.

^ Azimuthal direction of case acceleration for upward trace deflection on accelerogram (degrees clockwise

from north)

*** Accelerograph operational but did not trigger.

6 Time span between the first and last peak greater than 0.10 g.

33
Figure 1. Strong-motion stations in the Imperial Valley, California.

115°45'W 115°30'W

i \ V v-n

34
Table A - STATION LIST

Station Identification^ Site Structure Instrument

No Name

(Data Source)

Coord geology type/size location(s

5054 Bonds Corner

Hwys 98 & 115

32.69

115.34

NW 1-story bldg Ground level

5049 Borrego Air Ranch

Borrego Springs

33.19

116.28

NW 1-story bldg Ground level

5060 Brawl ey Airport

Brawl ey

32.99

115.51

NW Inst shltr Ground level

5073 Cabazon

Post Office

33.92

116.78

NW 1-story bldg Ground 1 evel

5053 Calexico Fire Station

Fifth & Mary

32.67

115.49

Nw 1-story bldg Ground level

5061 Cal i patri a Fi re Sta 33.13

115.52

NW 2-story bldg Ground level

5063 Coachella Canal Sta 1 33.64

116.08

N

W

1 -story bldg Ground level

5064 Coachella Canal Sta 2 Demosser

33.56

115.95

NW 1-story bldg Ground level

5065 Coachella Canal Sta 3 Siphon 24

33.51

115.77

NW 1-story bldg Ground 1 evel

5066 Coachella Canal Sta 4 Siphon 15

33.36

115.59

NW 1-story bldg Ground level

5046 Collins Valley 33.42

116.47

NW Inst shltr Ground level

5165 E C Differential Array

Dogwood Rd 32.80

115.54

NW Inst shltr Ground level

6 instrument

464 El Centro

Meadows Union School

32.80

115.47

NW Alluvium, more

than 300m

1-story bldg Ground level

5056 El Centro Sta 1 Borchard Ranches

32.96

115.32

NW Inst shltr Ground level

5115 El Centro Sta 2 Keystone Rd 32.92

115o37

NW

Inst shltr Ground level

5057 El Centro Sta 3 Pine Union School

32.89

TIC 0 O

1 1 b. Jo

NW 1-story bldg Ground 1 evel

955 El Centro Sta 4 2905 Anderson Rd 32.86

115.43

NW Alluvium, more

than 300m

Inst shltr Ground 1 evel

952 El Centro Sta 5 2801 James Rd 32.86

115.47

NW Alluvium, more

than 300m

Inst shltr Ground level

942 El Centro Sta 6 Huston Rd

32.84

115.49

NW Alluvium, more

than 300m

Inst shltr Ground level

5028 El Centro Sta 7 Imperial Valley Col

32.83

115.50

w

w 1-story bldg Ground level

958 £ 1 ucil Li U jLq 0 95 E Cruickshank Rd JL.U 1 115.53

Nu Alluvium, more

than 300m

Inst shltr Ground level

117* El Centro Sta 9 302 Commercial Av 32.79

115.55

NW Alluvium, more

than 300m

2-story bldg Ground level

35
Table A - STATION LIST (cont)

Station Identification^ Site Structure Instrument

No Name

(Data Source)

Coord geol ogy type/si ze location(s)

412 t i tentro o ta i

u

Community Hospital

32.78 N 115.57 W Alluvium, more

than 300m

1-story bldg Ground level

5058 L 1 ten tro sta 1 1 McCabe School

32.75

115.59

NW 1-story bldg Ground level

931 El Centro Sta 12 907 Brockman Rd 32.72

115.64

NW Inst shltr Ground level

5059 El Centro Sta 13 Strobel Residence

32.71

115.68

NW 1-story bldg Ground level

817 Hines Pumping Plant 33.71

115.63

NW Gatehouse Mid level

5055 Hoi tvi 1 le Post Office32.81

115.38

NW 1-story bldg Ground level

5067 Indio

So Cal if Gas Co 33.75

116,21

NW 1-story bldg Ground level

5050 ucot i mo we I i s Burro Bend Cafe

33.14

116.13

NW 1-story bldg Ground level

5051 Dsyarhii'f'Q Tact ^ i ^ a 32.93

115.70

NW 1-story bldg Ground level

5044 r i nun riaL uubc' vatury 33.61

116.46

N Inst shltr Tunnel

5C52 Plaster Ci ty

Storehouse

32.79

i u,oo

N

l/V 1-story bldg Ground level

5047 Rancho de Anza

Anza Borrego Park

33.35

116.40

NW Inst shltr Ground level

5062 Sal ton Sea Wildl ife Refuge

33.18

115.62

uw 1-story bldg Ground level

2S6 Superstition Mtn USAF Camera site

32.95

11 5.82

Nw Grani te 1-story bldg Ground level

5045 Terwi 1 1 i ger Val 1 ey Snodgrass Residence

33.48

116.59

NW Inst shltr Ground level

2316 Yuma Strand Ave 32.73

114.70

uw Inst shl tr Ground level

Ref: Western Hemisphere Accel erograph Station List; USGS, Open-File Rpt. 77-374.

* 1940 Strong-motion accelerogram recorded at this station. Ref: U.S. Eqks, 1940.

36
Table B - GROUND MOTION DATA

Station Identification1 Epicentral^

Distance

(km)

S - t 3

i WWVB 4 Interval | trigger

(sec) I time

Acceleration Duration^

> 0.1 g (sec) No. Name

(Data Source) Coord

Azimuth

(note 5)

Maximum

(g)

5028 El Centro Sta 7 32.83 W 26

| 4.6 | ** 230 0. 52 4.9

Imperial Valley Col 115 .50 w [1]

j up 0.65 5.5

(USGS

)

|

140 0. 36 3. 7

942 El Centro Sta 6 32.84 N 27 5* | 17:01 230 0. 45 7. 9 Huston Rd 115.49 W [1]

j up 1. 74 6. 2 (USGS )

|

140 0. 72 11.8

5054 Bonds Corner 32.69 N 6 2.4 j 16:57 230 0.81 13.2

Hwys 98 & 115 1 15. 34 W [3] up 0.47 12.0

(USGS )

I

140 0.66 13.3

958 El Centro Sta 8 32.81 N 27 5* | 17:00 230 0. 50 6. 9 95 E Cruickshank 115.53 W [4]

| up 0.55 5.8

(USGS

)

I

140 0.64 6. 9

952 El Centro Sta 5 32.86 N 28 5.1 | 17:01 230 0.40 7.6

2801 James Rd 115. 47 W [4] up 0. 71 5.6

(USGS) |

1

140 0.56 7.4

5165 E C Differential Array I 32.80 N 26 5 ** 360 0.51 | 10. 2 Dogwood Rd 115. 34 W [5] up 0. 93 | 7.0

(USGS) |

1

270 0.37 | 7.0 |

117 El Centro Sta 9 32. 79 N 26 - 360 0.40 | 7.4

302 Commercial Av 115. 55 W | [6] dwn 0.38 | 4. 7 (USGS) 1

1 090 0. 27 j 7.0 j

955 El Centro Sta 4 | 32.86 N 26 | 4.8* | 17:01 230 0.38 | 6. 5 2905 Anderson Rd 115. 43 W [7] up 0.32 | 6. 7 (USGS) 1

1

|

140 0.61 j 6. 8 j

5060 Brawley Airport 32.99 N 42 6.3 j 17:03 j 315 0.22 | 2. 2 | Brawley 115.51 w I [7] up 0. 18 | 5.2

(USGS) 1

1

1

225 0. 17 j 1.8 |

5055 Holtville 32.81 N I 19 4.1 | ** | 315 0.22 | 7.5

Post Office 1 15. 38 w I [8] up 0.31 | 7.0

(USGS) 1

1

1

225 0. 26 | 6.2

412 El Centro Sta 10 32.78 N I 27 4.9 | ** j 050 0.20 | 5.2

Community Hospital 115.57 W 1 [9]

| up 0. 15 | 2.2

fUSGS")

1

1 320 0.23 j 5. 1 j

5053 Calexico Fire Station | 32.67 N | 15 3.2 | 16:59 | 315 0. 22 | 9. 5 Fifth & Mary 115.49 w 1 [11]

!

1

up 0.21 | 8. 8 | (USGS) 1 225 0.28 | 10.8

5058 El Centro Sta 11 32.75 N | 27 | 5.6 | 17:00 | 230 0.38 | 6. 5 McCabe School 115.59 w 1 [13] up 0. 16 | 7. 7 j (USGS) 140 0.38 | 7.0

37
Table B - GROUND MOTION DATA (Continued)

Station Identification^ Epicentral^

Distance

(km)

S - t3 Interval

(sec)

WWVB 4 trigger

time

Acceleration Duration^

> 0.1 g (sec)

1« U» | Name

C oord

Azimuth

(note 5)

Maximum

(g)

5057 | El Centro Sta 3 j 32.89 N 28 5.4 ** 230 0.22 6.2

Pine Union School 115.38 W [13] up 0. 15 5.6

(USGS ) 140 0. 27 6. 0

5051 Parachute Test Site 32.93 N 47 7.0 ** 315 0.20 1.5

(USGS) I 115.70 W [15] up 0. 18 5.2

I

225 0. 11 1.4

5115 El Centro Sta 2 32.92 N 31 6 17:01 230 0.43 5.7

Keystone Rd 115. 37 W [16] up 0. 17 9. 3 (USGS) 1 140 0.33 9. 2

931 El Centro Sta 12 32.72 N 30 5. 2 17: 01 230 0. 11 4.9

907 Brockman Rd 115.64 W [18] up 0.08

(USGS) 1 140 0. 15 3.8

5061 Calipatria Fire Sta 30.13 N 57 7.4 17: 06 315 0.09 - (USGS) I 115. 52 W [21] up 0.07

I

225 0. 13 1 peak

5059 El Centro Sta 13 32. 71 N 34 5.

17: 02 230 0. 15 5.0

Strobel Residence 115.68 W [22] up 0.06

(USGS) I 140 0. 12 2. 4

5056 El Centro Sta 1 32.96 N 37 6 17: 02 230 0. 15 3. 1 Borchard Ranches 115. 32 W [22] up 0. 10 1 peak

(USGS

)

140 0. 15 4. 8

286 Superstition Mtn 32.95 N 57 7.2* 17:05 135 0.21 1. 1 USAF Camera site 115.82 W [26] up 0.09

(USGS) I 045 0. 12 0. 6

5062 Salton Sea 33. 18 N 66 3.5* 17:11 315 0. 10 1 peak

Wildlife Refuge 115.62 W [28] up 0.06

USGS ) 225 0. 13 1 p eak

5052 Plaster City I 32. 79 N 52 5* ** 135 0.07 - Storehouse 115.86 W [31] up 0.03 - (USGS) j

1 045 0.05 -

5066 Coachella Canal Sta 4 | 33.36 N 84 8.5 17: 11 135 0. 14 0.5

Siphon 15 115.59 W [47] up 0.04

(USGS) I 045 0. 11 0.3

5050 Ocotillo Wells 33. 14 N 93 7.5* 17: 15 315 0.05

Burro Bend Cafe | 116. 13 W [59] up 0.03

(USGS) 1 225 0.04

2316 Yuma 32.73 N OU * 17:11 090 0.03 | Strand Ave 114.70 W [61] up 0.02

(USBR/USGS) | 360 0.03 |

$8
Table B - GROUND MOTION DATA (Continued)

Station Identification^- Epicentral

Di s tance

S - t3 Interval

WWVB 4 trigger

Acceleration Duration^ | > 0.1 g |

No. Name

(Data Source) C oord

(km) (sec) time Azimuth Maximum

(note 5) (g)

(sec)

5065 Coachella Canal Sta 3 | Siphon 24

(USGS)

33.51

115. 77 NW

105

[67]

***

5049 Borrego Air Ranch

Borrego Springs

(USGS)

33. 19

116. 28

NW

108

[74]

8.2 17: 20 315

up

225

0.04

0.02

0.03 -

5064 Coachella Canal Sta 2 | Demosser

(USGS)

33.56

115. 95 NW

117

[79]

***

5047 Rancho de Anza

Anza Borrego Park

(USGS

)

33.35

116.40 w

127

[92]

* 17: 27 135up 045

0.03

0. 02

0.02

5063 Coachella Canal Sta 1 (USGS)

33.64

116.08

NW

131

[92]

* 17: 24 135

up

045

0. 02

0.02

0.03 j

5067 Indio

So Calif Gas Co (USGS)

33.75

116.21

NW

148

[109 j

***

5073 Cabazon

Post Office

(USGS)

33. 92

116. 78

NW

196

[158]

17:55 270

up

180

0.01

0.02

0.02 |

Ref: Western Hemisphere Strong Motion Accelerograph Station List - 1976; USGS, Open-File Report 77-374.

Distance from epicenter at 32.64 N and 115.33 W. Bracketed number is distance to the nearest point on the 1940 Imperial Fault trace. Reference: Sharp, R. , 1977; Open File Report 77-815; Misc Field Studies

Map MF 838.

S-wave minus trigger time.

* S - t is questionable.

Trigger time in minutes and seconds after 288 days (UTC) as determined from WWVB time time code.

Millisecond accuracy is possible.

** WWVB time code not legible.

Azimuthal direction of case acceleration for upward trace deflection on accelerogram (degrees clockwise

f rom north)

*** Accelerograph operational but did not trigger.

Time span between the first and last peak greater than 0.10 g.

39
THREE-DIMENSIONAL SYNTHETIC ACCELEROGRAMS

FOR DESIGN EARTHQUAKE GROUND MOTIONS

by

Makoto Watabe

ABSTRACT

Three-dimensional earthquake ground motions are simulated by the use of the principal

axes concept [1]. Two major procedures for mathematical modeling which simulate the

three-dimensional earthquake ground motions are introduced.

The first is a completely synthetic approach, simulated by the product of a stationary

random process and a deterministic intensity function in each component independent of one

another. The relation between maximum response and the factors of simulated ground motions

are theoretically established, and the relation is utilized for the simulation of

earthquake ground motions having a pre-established response spectra.

In the second procedure, it is found that the phase angles of recorded accelerograms

play an important role in determining the envelope of time history, by generating a "phase

wave." The role of phase angles on the covariances of three-dimensional earthquake ground

motions is also examined. Based upon the results, a set of phase angles of the three components of ground motion from a real earthquake are applied to simulate three-dimensional

earthquake ground motions along principal axes, appropriate to the pre-established response

spectra in each component.

40
INTRODUCT ION

Up to the present (1980) , in most cases of dynamic analyses of structural systems

subjected to strong earthquake ground motions, one component of earthquake motions alone

has been considered. The consideration of simultaneous response effects due to three

translational components of earthquake ground motions, however, should not be neglected in

the seismic design of structures, such as nuclear power plants, in which a detailed and

precise seismic analysis is required.

If a very large number of recorded ground motion accelerograms were available,

representative stochastic models could be established directly by statistical analyses

based on these data. Unfortunately, recorded strong motion accelerograms are quite limited.

Therefore, one is forced to hypothesize a model for generating simulated three-dimensional

earthquake ground motions. For one such model, three-dimensional earthquake ground motions

along "principal axes" have been proposed by Penzien and Watabe [1]. In these principal

axes, the corresponding variances of motions have maximum, minimum, and intermediate values

with zero co-variances. Utilizing this concept, three-dimensional earthquake ground

motions can be simulated along three-orthogonal axes.

A general form of one-dimensional simulated ground motion can be expressed as a

function of time t, such as:

a(t) = s(t) b(t) (1)

where a(t), c(t) and b(t) represent the simulated nonstationary ground motion, the

deterministic intensity function, and the stationary random motion, respectively. Threedimensional simulated ground motions along principal axes may be represented through the

relation

ax(t) = Cx(t)bx(t), ay(t) = £y(t)by(t) . az(t) = ^z(t) bz(t) (2)

where ax(t), ay(t) and az(t) are three components of ground motions simulated along three

orthogonal coordinates and bx(t), by(t) and bz(t) are stationary random motions

stochastically uncorrelated with each other.

In this paper, spectral characteristics of simulated ground motions are represented by

a pre-established response spectra. Therefore, the relation between the maximum response

41
values and the factors of the input process should be first established. A simple approach

to estimate the maximum response subjected to nonstationary random excitation is proposed

as a product of the square root of the mean square of response values and a magnification

coefficient, with the constant "R" expressing the nonstationarity of the input process.

In a stochastic modeling of earthquake ground motions, such as Eq. (1), phase angles

are generally considered as random values with a uniform probability distribution. However, it was found that phase spectra are important factors as well as amplitude spectra

in the time domain of each sample [9]. Based on this, the effects of phase angles on the

nonstationarity of time history and the covariances of three-dimensional earthquake ground

motions are examined. It is one of the purposes of this paper to utilize the phase spectra

of real accelerograms to simulate three-dimensional earthquake ground motions on the basis

of these basic studies.

Relation between maximum response values and input process

The estimation procedure of the maximum of stationary random vibrations of the

Gaussian process and its response has been studied by several investigators [2,3,4]. To

predict the maximum response subjected to a non-stationary random ground motion with an

arbitrary deterministic intensity function, a simple approach is here proposed. It is

defined by the relation:

where E denotes ensemble average, J(t) are response values in terms of t, and J^(t) and M

are the mean square of J(t) and the magnification coefficient of E [|j(t)| max] to

The equation of motion of a single-degree of freedom system excited by earthquake

ground motions a(t) can be expressed:

SIMULATION OF THREE-DIMENSIONAL EARTHQUAKE GROUND

MOTIONS USING A COMPLETELY SYNTHETIC PROCEDURE

E [| J(t) I max] (3)

x(t) + 25Xx(t) + X

2x(t) = -a(t) (4)

42
where x(t) is a relative displacement in the system having a critical damping ratio (£) and

an undamped natural circular frequency (A). The solution of Eq. (4) is obtained by using the

Duhamel integral (with zero initial conditions):

t J(t) = / - h-j(t - t) a (x) dx (5)

o

J

in which J(t) implies a relative displacement, a relative velocity or an absolute

acceleration of responses, hj(t) is a unit impulse response function and t is a dummy

variable

.

If a(t) is a nonstationary random process given by the product of a deterministic

intensity function £(t) varying slowly in terms of t and a stationary random process b(t)

having zero mean value, such as expressed by Eq. (1), the ensemble average of the mean

square of J(t) in the duration time Td is approximated in the form [5]:

E [J 2 (t)j = c 2 (t) E [J 2

( t )] (6)

where £ (t) is the mean square of the deterministic intensity function £(t) and -Jg(t) is

the mean square of responses Js (t) subjected only to the stationary ground motion b(t).

The stationary random process b(t) is assumed to be represented by the following

relation:

N

b(t) =

I Aj • cos(ojjt + <|>j) (7)

3=1

where N is the number of harmonic waves superposed, Aj is an amplitude at frequency uj and

c[>j is the random phase angle with a uniform distribution of probability in the range

between 0 and 2tt. Therefore, through the integration of square values of J s (t) over the

entire duration E [J 2 (t)] can be obtained as follows:

[J 2 (t)] = I |a? • iHjCi.^)! 2

. y (8)

J-l 2

43
in which Hj(i*ioj) is the complex frequency response function of J(t). X in Eq. (8)

represents the effects of transient state of response, that is 1 for wj # X and for

ujj = X is obtained as follows, by neglecting all the terms of the second order of E, and the

terms of the first order of £ less effective than the other terms.

XOj = X) = 1 - [1 - exp(-?XTd)] • [3 - exp(-5XTd)] / 2£XTd (9)

Next, considering past research and the results of numerical experiments such as shown

in Figure 1, the magnification coefficient M can be assumed by the following equation:

log M = f

l + f2

/ f

3 [(log Neq ) f 3 - 1] (10)

where Neq indicates the equivalent number of waves crossing the zero baseline to either

positive or negative direction over the entire duration. It can be estimated by taking the

spectral moment of response Js (t):

N

eq

= Td/2 n A [Jg(t)] / E [J 2 (t)J (11)

where the dot represents differentiation with respect to time t. The coefficients f\, f 2

and f3 in Eq. (10) can be found to be the function of a damping ratio £ and a nonstationarity of a(t), which is represented by the ratio of total power of a(t) to the one of the

stationary wave expressed as |c(t)| max x b(t) as ensemble average. This nonstationarity constant R is expressed as follows:

R = /t. E [a 2 (t)] dt / Td«|c(t)| 2 max • E[b 2 (t)]

= C

2 (t)/U(t)|2max (12)

Through the rearrangement of the results of numerical experiments such as shown in

Figure 2, and the trial and error procedure, the coefficients f^, f 2 , and f3 become:

f

1 = 0.2 + 0.055 logR + 0.022R -0 ' 86

£ 0 - 076 R

-1 - 3

0.12 r

-0 - 96

f [l - exp(-100C)] ( (13)

2

f

3 = R

115 [0.28 + 0.84 exp(-13£)]

The resulting magnification coefficients are also presented in the solid lines of Figures 1

and 2.

44
The relation between maximum response and amplitude Aj in Eq. (7) can be established,

by substituting Eqs. (6) and (10) into Eq. (3). From the results verified by the computer

simulation, it has been found that the maximum response subjected to a nonstationary ground

motion can be predicted on the basis of this procedure within a 5 percent error range [5].

An example of simulation

An example of three dimensional earthquake ground motions simulated along principal

axes is here introduced, in which these three components have preestablished response spectra in each component. In this procedure the amplitude spectra of stationary random motion

b(t) represented by Eq. (7) are estimated as shown in the Figure 3 block chart, applying the

method described above.

For this example, the parameters proposed in [7] are applied, where the velocity

response spectra Sv (in cm/sec) in the horizontal component are:

log S v (m, x, T, O = 0.607 m - 1.19 log x - 1.15 + g\ (m, T) -

g 2 (x, T) + g 3 U, T) (14)

where m, x, T and E, indicate magnitude of earthquake, hypocentral distance (in km), period

(in sec.) and critical damping ratio, respectively. The function gi(ra, T), g 2 (x, T) and

g3(£, T) in Eq. (14) are defined as follows:

gl (m, T) = a

x x [1 - (T/a

2

) 33

] x [1 + T exp(l - a

4T)]

where, aj = 0.015/(m/8) 14 + 0.055, a

2 = 0.045 x 1.6m

a

3 = 1.8 x (m/8) 13

/ [(m/8) 13 + 0.15] , a

4 = 0. l/(m/8) 15 + 0 .

9

g 2 (x, t) = 0.25 x X

0,1

x (1/ /T~-l)

and, g 3 (5, T) = I- log [[log

e

2. 72( 4irn5 / 1 . 78 + 1)] x [1 - exp(-4imO] / 4irn5 ]

where, n = T 0 /T, log T 0 = 0. 31 m - 1 .

2

The velocity response spectra defined by Eqs. (14) and (15) are derived from the

results obtained through statistical analysis using 74 accelerograms recorded on rock

sites in Japan [6]. The dotted lines in Reference-Figure 1 are some examples of velocity

response spectra with the constant hypocentral distance 50 km for the various values of

magnitude, obtained by Eqs. (14) and (15). Similar examples are shown in Reference-Figure

45
2 with the constant magnitude 8 for the various values of hypocentral distance. The solid

lines in Reference-Figures 1 and 2 indicate the velocity response spectra obtained using

the results of statistical analysis mentioned above.

The contents of this example are assumed as follows:

(i) The magnitude of earthquake is 8 and its epicentral distance is 20 km.

(ii) Based upon the discussions in [1] and [6], it is assumed that the major and intermediate principal axes are horizontal and the minor principal axis compose the vertical

components of ground motions.

(iii) The velocity response spectra with 5 percent damping by Eq, (14) are converted to

pre-established response spectra in the major and intermediate principal axes by multiplying by 1.14 and 0.86, respectively. The above values of 1.14 and 0.86 are derived from

the assumption that the energy content of earthquake ground motions can be represented by

the variance of acceleration [8] , the square root value of the variance can be identical

to the spectral intensity, and the fact that from the results of analysis in [6] of three

dimensional earthquake ground motion observations, the average ratio of the variance in

the intermediate component to the one in the major component is about 0.55.

(iv) As for the minor component, the mean ratios of the response spectra in [7] to

convert horizontal components into verticals are utilized.

(v) The shapes of deterministic intensity functions are assumed to be the ones proposed

in each horizontal and vertical components in [7], Assuming that the total time of these

proposed deterministic intensity functions is 70 sec, time histories of duration 35 sec

duration are generated.

Across the ensemble, only one sample is presented here. The time histories of

simulated accelerograms and their velocity response spectra compared to the pre-established

response spectra are shown in Figures 5 and 6, respectively. The maxima of each component

and the direction of principal axes of generated three components over the entire duration

are presented in Table 1. The directions are represented by the declination \\i from the

pre-assumed minor axis (z) and the direction 9 from the pre-assumed major axis (x) on the

x - y plane as shown in Figure 4. From these results, it will be seen that the calculated

response spectra agree well with the pre-established ones and the direction of each

component is close to the principal axis.

46
SIMULATION OF THREE-DIMENSIONAL EARTHQUAKE GROUND MOTIONS

INCLUDING PHASE ANGLES OF REAL ACCELEROGRAMS

Characteristics of phase angles

Most of the recent research in the field of earthquake ground motion simulation

suggests the randomness and the uniform probability distribution of phase angles in earthquake ground motions. It is also a fact, however, that the phase angle spectra of real

accelerograms play an important role in determining the nonstationarity of the time history. From the results of analysis on the roles of phase angles in earthquake ground

motions, it has been found that phase angles are strongly correlated to deterministic

intensity functions and the shapes of response spectra [9,10], Therefore, it may be wise

to utilize positively the properties of phase angles to control envelopes of simulated

accelerograms rather than to apply the synthetic random phase angles with a uniform probability distribution which may contribute little to the envelope of accelerograms. Using a

set of phase angles from a real earthquake accelerogram, independent of amplitude characteristics, may be the proper procedure to realize the above idea.

Figure 7 shows the original accelerogram of Taft 1952, EW component, and a simulated

one which has uniform amplitudes in each frequency component with the same set of phase

angles of the original accelerograms. Let this kind of accelerogram be called as "phase

wave." The phase wave is then generated in the following manner: A digitized accelerogram

ar (t) due to real earthquake can be expanded into a Fourier series as follows:

N a_(t) = I A. • cos(u).t + tf) (16)

J=0

Phase wave aD (t) in Figure 7 is generated using the relation:

a (t) = ?l cos((d.t +<t>i P j=l J f) (17)

where P is the scaling factor of intensity, and the high frequency components in a r (t) are

excluded in ap(t). It is clearly shown in Figure 7 that the envelope of the phase wave is

quite similar to the original one. Therefore, this result suggests that one can simulate

47
the wave with arbitrary spectral amplitudes, having a similar envelope to the original one,

provided that the distribution of spectral amplitudes is rather uniformly and widely

scattered

.

It has been pointed out by the study of Ohsaki, et al., [9] that the characteristic

of nonstationarity of an earthquake ground motion can be expressed by the distribution of

phase differences A<J>, that is given by

A^ = <j)j + i - 4>j (18)

where A<j>j is defined in the range - 2ir<A<j><0, and <f>j is the phase angle at frequency ojj.

Figures 8 and 9 show the relative frequency of phase angles and phase differences of Taft

1952-EW component in the 30 sec time duration, where high frequency components are

excluded. It is seen from these figures that the distribution of phase angles is nearly

uniform, while the probability distribution of phase differences has some characteristics

similar to that of Gaussian. As pointed out in [9], the distribution of phase differences

normalized by -2 uhas a quite similar feature to the envelope of time history normalized by

the duration. These results suggest that it is possible to generate the nonstationary

earthquake ground motions without applying deterministic intensity functions by using

these characteristics of the distribution of phase differences.

For the simulation of components along principal axes, these phase angles of real

accelerograms can be also applied. Provided that two components a^Ct) and a^(t) are the

components of three dimensional accelerograms along principal axes, the covariance between

these components can be expressed as follows:

N

a k (t) • 3jl (t) = j I A

kj • A

£j

• cos(<}>

k;j - <j) £j

) = 0, k * I (19)

j=0

The resulting equation suggests that the covariances of three dimensional

accelerograms phase angles play important roles as well. Considering the fact that, in

most cases, effective frequency domain in both real accelerograms and simulated ones is

similar and the resultant variances defined by three-dimensional components produce a

rather smooth function in terms of transforming vectors, it can be assumed that the

simulated components using the phase angles of real accelerograms transformed into the

48
principal axes are already closely oriented to principal axes. The term "original" in

Table 2 means the directions of principal axes of the three dimensional accelerograms

Taft-1952, and El Centro-1940 from the recorded three orthogonal axes, in which the x-axis

in Figure 4 corresponds to the axis of the NS component. The term "phase" in Table 2 mean

the directions of principal axes of simulated three dimensional phase waves with the

phase angles of real accelerograms transformed into principal axes, from simulated three

orthogonal axes. In these phase waves, it is assumed that the variances of intermediate

and minor components to the major component variance are 0.55 and 0.18, respectively [6].

It will be recognized in Table 2 that the directions in phase waves are very close to

those of principal axes. This fact may support advantages of this procedure to generate

the simulated threedimensional ground motions along principal axes.

An example of simulation

Based on the discussion above, it is possible as an example to generate simulated

three-dimensional earthquake ground motions with pre-established response spectra through

an iterative procedure. This example has the phase angles of Taft-1952 accelerograms

transformed into the principal axes. The earthquake parameters for the pre-established

response spectra are the same as those mentioned in the previous example. The results are

shown in Figures 10 and 11. The resultant direction of principal axis and the maxima of

simulated ground motions are presented in Table 3. It can be seen from these results that

the simulated ground motion fit the pre-established response spectra very well and the

direction each component is close to the principal axis. In addition, the wave forms of

the simulated time histories are similar to the original ones.

CONCLUSIONS

Two major procedures for the simulation of three-dimensional earthquake ground motions

which utilize the concept of the principal axes of earthquake ground motions have been

introduced. One is a completely synthetic way as represented by Eq. (2), and in the other

the phase angles of real accelerograms which are transformed into principal axes are

utilized.

49
The relation between the maximum response and the input process factors in the former

method has been established because of the importance of response spectra in engineering.

This relation also can be used to predict the maximum response of a multi-storied structure and their floor responses as ensemble average [5].

The effects of phase angles on the envelope of time history and the covariances of

three dimensional earthquake ground motions in the second procedure have been examined.

It has been found that one can generate simulated three-dimensional earthquake ground

motions along principal axes with wave forms similar to real accelerograms, using the

actual sets of phase angles. On the basis of these examinations, it is concluded that

for practical purposes in the seismic design of structures, the effect from vertical

earthquake ground motions can not be neglected and the simulated three-dimensional earthquake ground motions calculated by the second procedure are recommended for use.

Note: A similar paper was submitted to 7th World Conference on Earthquake Engineering

held in Turkey in 1980.

REFERENCES

[1] Penzien, J. and Watabe , M. , 1975, "Simulation of Three-Dimensional Earthquake Ground

Motions," Int. J. Earthq. Engng. Struct. Dyn., Vol. 3, pp. 365-373.

[2] Davenport, A. G., 1964, "Note on the Distribution of the Largest Value of a Random

Function with Application to Gust Loading," Proc . Inst. Civ. Engng., Vol. 28,

pp. 187-196.

[3] Vanmarcke, E. H., 1976, "Structural Response to Earthquakes," in Seismic Risk and

Engineering Decisions, ed . C. Lomnitz and E. Rosenblueth, pp. 287-337.

[4] Iyengar, R. N. and Iyengar, S. R., 1970, "Probabilistic Response Analysis to Earthquake," ASCE, EM 3, Vol. 96, pp. 207-225.

[5] Watabe, M. and Tohdo , M. , 1980, "Research on the Simulation of Three-Dimensional

Earthquake Ground Motions, PART-II," Proc. 5-JEES, Tokyo, pp. 113-120.

[6] Watabe, M. and Tohdo, M., 1979, "Analyses on Various Parameters for the Simulation

of Three-Dimensional Earthquake Ground Motions," Proc. 5-SMIRT, Kl , Berlin, pp. 1-11.

[7] Ohsaki, Y., Watabe, M. and Tohdo, M., 1980, "Analyses on Seismic Ground Motion

Parameters including Vertical Components," submitted to 7-WCEE.

[8] Arias, A., 1970, "A Measure of Earthquake Intensity," in Seismic Design for Nuclear

Power Plants, ed . R. Hanson, MIT Press, pp. 438-483.

[9] Ohsaki, Y . , Iwasaki, R. , Ohsawa, I. and Masao, T., 1978, "A Study on Phase

Characteristics of Earthquake Motions and their Application," Proc, 5-JEES, Tokyo,

pp. 201-208.

[10] Kubo, T. and Suzuki, N. , 1978, "An Application of Synthetic Earthquake Ground Motions

to Response Analysis," Proc. 5-JEES, Tokyo, pp. 89-96.

50
-

LOG. EQUIVALENT WAVE NUMBER Fig.l Relation between Magnification

Coefficient and Equivalent Wave

Number of Response

2 o x • Experiment u

- (0.10, 0.05, 0.02 Dampings) ; J

0.2I — 1 1 1 -0.6 -n.A -0.2 0. LOG. NONSTATIONARITY Fig. 2 Relation between Magnification

Coefficient and Nonstationarity of

Input with the Parameter of Dampings,

at Neq=10

] start!

INPUT GIVEN SPECTRA : 0Sj DURATION TIME : Td NONSTATIONARITY : R

SETTING AMPLITUDES TO 1 : A=1

CALCULATION OF MAGNIFICATION

COEFFICIENT AND MEAN SQUARE

: M , E[j~]

I CALCULATION OF INITIAL AMPLITUDES

: Ai=Sj;/[Mi*/ETn']] , i=1,N

CALCULATION OF Mi and E[5t]

CALCULATION OF Sjj (=Mt*/j[j7

CHANGING AMPLITUDE TO FITSj

:Ai=Ai*Sj'i/Sji

Fig. 3 Block Chart for Estimation of

Amplitude Spectra of Input

Fig.U Principal Axis in

Three Orthogonal Coordinate

System

51
SAMPLE SIMULATED GROUND MOTIONS

MINOR COMPONENT

«/vv—1

S 300. I NTERMED I PTE COMPONENT

a:

o

o

en

300.

300. MAJOR COMPONENT

Jin. fell UUW Mjfl 1 ^UU.llU./l, .Hm.LA A».

10. 20. TIME C SEC )30.

Fig, 5 Time Histories of Sample Simulated Accelerograms

Ta"ble 1 Maxima and Directions

of A Sample Simulated Ground

Motion

Component Major Inter. Minor

Estimated A 230. 172. 96. Maxima V 38.5 28.9 18.6

Simulated A 253. 144. 82. Maxima V 52.6 33.6 16.0

Direction * 88. 90. 2.

9 - 8. 81. 24. A ; Acceleration (gal) V ; Velocity (kine)

,9 ; degree

RE6P0N6E SPECTRA WITH 57- DAMPING

T

0. 10 1.00 PERIODtSEC ) Fig. 6 Pre-established Response Spectra

and Calculated Ones by Sample Simulated

Accelerograms

52
5.0

0.0 Phase (rad.) 2ti Pig. 8 Relative Frequency of Phase

of Taft 1952, EW Component

T

Phase Difference (rad.l -2n Fig. 9 Relative Frequency of Phase

Difference of Taft 1952, EW Component

Table 2 Directions of Principal

Axes of Original Waves and

Phase Waves

Record Component

Original Phase

* 8 e

Taft

-1952

Major 73. 57. 88. - 3. Inter. 86. -32. 88. 87. Minor 18. 71. 3. 45.

El Centro

-1940

Major 87. -25. 88. 9. Inter. 89. 65. 89. -81. Minor 4. -43. 2. - 2. (degree) 53
FITTED SIMULATED GROUND MOTIONS

MINOR COMPONENT

INTERMEDIATE COMPONENT

300.

i 1 TIME ( SEC ) Fig. 10 Time Histories of Simulated Accelerograms

by Use of Phase Angles of Taft 1952

RE6P0N6E SPECTRA WITH 5* DAMPJNG

t- i.o

i.

A, jf^7* Major

Inter

* Minor

Pre-established Calculated

o. to 1.00 PERiODC SEC 3 Fig. 11 Pre-established Response Spectra

and Calculated Ones by Simulated Accelerograms through Iterative Procedure

Table 3 Maxima and Directions

of Simulated Ground Motions

by Use of Taft-1952

Component Major Inter. Minor

MaxijTB A 313. 172. 82. V 65.0 30.2 19.3

Direction

. 86. 85. 6.

e 3. -87. 55. A ; Acceleration (gal) V ; Velocity (kine)

,8 ; degree

54
RESPONSE SPECTRA ON BED ROCK

Danping Ratio 0 05 Magnitude

by A,B,C

Modified

ij

Tpr-

/ /

——

7

if J'

f

• ^><^^ 5

Reference-Figure 1 Velocity Response Spectra

on Rock Site at Hypocentral Distance 50km

0.10 1.00 PERI0DC SEC ) Reference-Figure 2 Velocity Response Spectra

on Rock Site at Magnitude 8

55
STUDY OF SHORT AND LONG PERIOD DYNAMIC GROUND

CHARACTERISTICS IN THE SENDAI DISTRICT

by

Yoshikazu Kitagawa

Makoto Watabe

ABSTRACT

The dynamic characteristics of soil-layers in a wide range of periods are investigated

using the Sendai district, damaged by the 1978 Off-Miyagi Prefecture Earthquake, as an

example. The deep-ground characteristics of long periods are estimated from the seismic

data published by J.M.A. To estimate short period earthquake motions on the ground surface, strong motions observed in a building are analyzed considering building-subsoil

interaction to produce the incident wave at the structural base rock. The seismic

microzoning maps are compared to the actual building damage, and a close correlation

between the two is recognized.

56
INTRODUCTION

One of the most important problems in the field of earthquake engineering is to

predict the maximum earthquake motions in a certain place which may be defined by the

seismic activity and ground characteristics. In order to clarify this more reliably, it

is necessary to investigate separately each of them. Generally the ground motion in the

frequency domain (G(jw)) will be expressed by the following equation,

G(jw) = [FS(jw) • TF(jw)] • ST(jw) , (1)

where FS(jw), TF(jw) and ST(jw) represent the spectrum of the focal mechanism, the overall

transmission function for the generation and propagation of earthquake waves, and the function for the wave propagation of local soil layers, respectively. The first parentheses of

the equation represents the theoretical seismic records based on the focal mechanism, and

the latter for the ground motion dealt with in the earthquake engineering.

The long period (10 > T > 1.0 sec) which corresponds to the ground characteristics of

a fairly large area is amplified by the soil layers of depth varying from a few hundred

meters to 1~2 kilometers and, the short period (T <_ 1.0 sec) corresponds to ground characteristics just beneath a construction which is affected by soil layers with a depth of

less than a few hundred meters. The former represents seismic macrozoning and the latter

microzoning. Using these two categories, a seismic base rock with a shear wave velocity

of 2~3 km/ sec, and a structural base rock with a shear wave velocity of 700 m/sec are

established

.

The estimates of ground characteristics in the short period range are based on the

results of the analytical investigation of the damped, elastic system which is independent

of frequency and microtremor measurements. The various seismic microzoning maps thus

obtained are compared to the distribution of real damage to buildings.

OUTLINE OF GEOLOGICAL CONDITION

Topographical Aspects

The geological conditions of surface soil in the Sendai district are generally

classified into the three areas shown in Figure 1: (1) the hilly tertiary terrain, (2) the

terrace area, and (3) the alluvial plain. The oblique NE-SW line passing near the center

57
of the map is called the Rifu-Nagamachi tectonic line. The area west of the tectonic line

is characterized by the hilly tertiary terrain and several levels of terraces. The surface

deposit on this terrace is loam underlain by hard clay, gravels, pelite and shale. The

hilly terrain is very hard andesite and shale, but the surface is covered with loam in several places. The aluvial plain develops east of this line and is mostly sand, silt, and

gravel.

The depth of the tertiary base rock varies erratically near the tectonic line.

Several areas in the plain are covered by very soft peat or mud. The soil profile and the

results of seismic refraction tests by the well shooting method at representative places

in the terrace and alluvial areas are shown in Figure 2.

Microtremor Measurements

The microtremors were measured at several sites shown in Figure 1 by an electromagnetic

seismometer, natural period of 1.0 sec, to get the dynamic properties of soil layers [1].

The area where the measurements were performed belong to two types of geological conditions (2) and (3) mentioned above. The representative Fourier spectra of microtremors in

NS and EW directions are shown in Figure 3, where sites G-06 through G-08 and sites G-17

and G-18 are on the outcrop of terrace and on the alluvial plain, respectively. Long period peaks of more than 1.0 sec reflect the deep ground characteristics, whereas short

period peaks of less than 1.0 sec reflect the shallow ground ones.

Quake Degree for Deep Ground Structure

The values of the maximum displacements of NS and EW components (Ajqg and Agy , micron)

observed at various stations of the Japanese Meteorological Agency are substituted in the

Tsuboi's formula to get the magnitude (M) of the earthquake as the average. Under the

assumption that 10^ is always regarded as the standard amplitude in the range of the long

period reflecting the deep ground characteristics of a fairly large area, the following

value is calculated at the J.M.A. station in the Sendai district:

m i=i

(2)

58
where 3 = M - 1.73 logA + 0.83. A, and m denote the epicentral distance (km), the

period of the phases giving the maximum displacement amplitude and the total number of the

earthquake data, respectively [2]. Consequently, the value of F means the averaged ampli—

tude at the Sendai station normalized by the standard one. Figures 4 and 5 show the F

value calculated by Eq. (2) and the number of data, respectively. It is found that the

degree of the quake relative to the standard one Is high at the 3.0 and 5.0 sec periods,

and that the data number is great at the 1.0 and 2.0 sec periods.

Setting Up of Structural Base Rock

The transfer function, which is defined by the ratio of the wave on the ground

surface (Uo) to the incident wave (Uin), for the deep underground structure and for the

shallow one at a representative place in the terrace near the Sendai station of J.M.A., is

shown in Figure 6. The shear wave velocities V and densities p of the various soil-layers

were determined by the seismic refraction test, the seismic data of strong motion seismographs of J.M.A., and the boring tests. As seen in Figure 6, there are some peaks which

vary with the levels of the base rock but the predominant frequency between 3.0 and 4.0 Hz

is common to both. This frequency also appears in the results obtained by microtremor

measurements at the sites G-06 through G-08

.

Consequently in this study, the structural base rock with shear wave velocity of

700 m/sec is set up to get the ground characteristics in the short period range of less

than 1.0 sec. The values of shear wave velocity at intervals of 0.5' in both latitude and

longitude in the Sendai district are estimated from the soil profile, results of the penetration tests, and the geological conditions. Figure 7 displays, as an example, the profile of shear wave velocity at section A-A 1 (line of N38°15'30") indicated in Figure 1.

SEISMIC MICR0Z0NING

Incident Wave Motions in the Structural Base Rock

In order to estimate the earthquake motions on the ground surface, the strong motions

observed at the Sendai Sumitomo Building [3] during the Off-Miyagi Prefecture Earthquake of

June, 1979 (M = 7.4) were analyzed with consideration of building-subsoil interaction to

get the incident wave at the structural base rock with shear wave velocity of 700 m/sec.

This is an eighteen-story, steel-framed, reinf orced-concrete structure with two-story
Fig.l Geological Condition

i t >

Fig. 3 Fourier Spectra

3"

>- 50

00 10 10 30 AO 50 SO

PEBIOO T (skI Fig. 4 F-Value Fig. 5 Number of Data

^2 V p H Q

(ms«e) r (m) u,

^H^yy—17—> ^—irv^—

a

.

-12-f» 18 70 200

40 13 CO 200

-28 523 20 60 400

703 20 (TU) 500 -KO

BOO 2J 8600 1000

3000 10 cd 2000 f

1 2 3 4 5 S • 1/n

*£-u; C=- Frea(Hz) /Q

(ml Fig. 7 Profile of Shear Wave Velocity (A-A' Section)

Fig. 6 Transfer Function

60
basements. The plan and elevation of this building are shown in Figure 8, where the

locations of strong motion accelerographs are indicated. Figure 9 shows the accelerograms

recorded at the 2nd basement floor. Figure 10 displays the Fourier spectra for the accelerograms at the 18th floor, and the 2nd basement floor together with the spectral ratios

between the both.

The building and subsoil were idealized as an elastic one-dimensional continuous media

with some amount of damping which is independent of frequency. The parameters required in

the analysis were determined from the results of the Fourier spectral ratios and seismic

prospecting. The rigidities and densities of the layers representing the building and the

first soil layer were modified so as to approximately satisfy the compatibility at the

boundary, assuming that shear waves are propagated and reflected only in the vertical

direction [4] . Figure 11 shows the observed and the calculated maximum accelerations in

NS and EW directions together with the assumed shear wave velocities of this particular

site. Figure 12 indicates the response acceleration spectra with damping ratio (h) of 0.05

for the observed wave at the 2nd basement floor, the calculated incident wave at structural

base rock, and the calculated wave at outcropping of the 2nd basement floor level in NS and

EW directions.

Using those two components at the base rock, a wave form for the major principal

axis [5] was calculated and taken as the final incident wave. The direction of its axis

was N18.5°E in two-dimensional space. Figure 13 shows the resultant incident wave motion

at the structural base rock.

Distribution of Dynamic Characteristics of Underground Structures

The predominant period and the magnification factor are calculated at intervals of

0.5' in both latitude and longitude using Haskell's Method. The distribution maps of the

predominant period and the magnification factor of the underground structure above the

structural base rock are shown in Figures 14 and 15, respectively. As found from these

figures, the predominant periods are around 0.2~0.3 sec in the terrace area, and 0.3~0.8

sec in the alluvial plain. Similarly, the magnification factors are about 2~3 in the former

and 3~5 in the latter.

61
It

C.L ~L

rr —

j vNN ;i (

' • :

Iv l>

i .:

! ' VAN

! ; '. b

ftMi

-400

60C

R»x. ACC. - 253 CXL

r

v NS

js.o :i.c CM.

fJ «v.f,V^..A^

Majl. Acc. - 227 CAi.

<

''' EW

Fig. 8 Plan and Section

10.0 15.0 20.0 25.0 JO.O Fig. 9 Accelerograms at 2nd Basement Floor

62
Distribution of Maximum Acceleration and Velocity

The maximum acceleration and velocity on the ground surface are calculated as the

response due to the aforementioned incident wave on the structural base rock. The distribution maps of maximum acceleration and velocity are illustrated in Figures 16 and 17,

where the values of contour lines are normalized by the maximum value of those maxima.

As seen in these figures, the value in the alluvial plain is about 1.5 times as much as

that of the several terrace levels.

COMPARISON OF STRUCTURAL DAMAGE WITH MICROZONING MAPS

Extents of Damages to Buildings

Figure 18 shows the distribution of damage to reinforced concrete buildings, steel

buildings, wooden houses, and landslide areas in the Sendai district. The severe damage

to reinforced concrete and to steel structures concentrated on the deluvium and the alluvial plain. Damage to wooden houses was on the alluvium and on the several terrace levels.

Three-story reinforced concrete buildings were extensively damaged by shear failure of

columns in the first story. Almost all of damage to low-rise steel structures was brought

about by buckling, bracing breaks, and anchoring failures. Damage to wooden houses is

classified into two types: (1) damage due to ground motions on the alluvium plain and

(2) damage due to terrace embankment landslides.

Comparison of Distribution of Damage with Seismic Zoning Maps

From a comparison of Figure 18 with Figure 16 or Figure 17, a close correlation

between the location of severely damaged structures and the distribution of the maximum

acceleration is generally apparent. It should be noted, however, that the soil is assumed

to behave elastically in the foregoing analyses. In the case of severe earthquake, this

will be a crude approximation and should be modified. Underground structures have been

modeled by the hysteretic nonlinear systems although reliable values about the nonlinearity

of soil in this district are not available. Figure 19 illustrates an example which gives

the distribution of maximum accelerations obtained by such inelastic analysis where the

Ramberg-Osgood hysteresis is assumed for the restoring force characteristics of soil

deposits [6]. It appears that maximum accelerations tend to distribute more uniformly

than those in Figure 16.
64
CONCLUDING REMARKS

The dynamic characteristics of underground-structures in the Sendai district have

been investigated. Deep ground characteristics were estimated from the J.M.A. strong

motion seismograph data, and the shallow ground characteristics were evaluated by microtremor measurements, boring data, penetration tests, and so on. The incident wave motion

was calculated using the strong motion accelerograms recorded during the Off-Miyagi Prefecture Earthquake. The predominant periods, magnification factors, maximum accelerations,

and velocities on the ground surface were computed, and the real damage to buildings compared with seismic raicrozoning maps. The close correlation between both is recognized.

As a result, it is pointed out that the method of analytical investigation presented here

is appropriate for seismic microzoning maps [7].

ACKNOWLEDGMENTS

The authors would like to acknowledge the continuing encouragement and discussion

of Dr. Y. Matsushima, Tsukuba University. In preparing the present paper, Mr. M. Tohdo,

Res. Memb. of Toda Const. Co. Ltd., contributed a great deal. The authors also wish to

express their sincere thanks to Dr. Y. Sugiraura, Head of Geotechnical Eng. Div., B.R.I.

,

for his help with a large amount of geological data; and to Mr. T. Kashima and

Miss M. Makishima, Membs. of B.R.I. , for drawing figures and typing the manuscript. This

work was partially supported by funding from the Science and Technology Agency.

REFERENCES

[1] "Report on the Damage by 1978 Off-Miyagi Prefecture Earthquake (in Japanese),"

Report of B.R.I. , No. 86, pp. 57-81, 1979.

[2] Kitagawa, Y. and Ozaki, M., "Study on Regional Characteristics of Maximum Earthquake

Motion in Japan," proc of 5-th J.E.E., pp. 1-8, 1978.

[3] "Report on the Damage by 1978 Off-Miyagi Prefecture Earthquake (in Japanese),"

Rept. of S.T.A., No. 15, pp. 58-60, 1978.

[4] Osawa, Y., Kitagawa, Y. and Ishida, K. , "Response Analyses of Earthquake Motions

Observed In and Around a Reinforced Concrete Building Including Building-Subsoil

System," Proc. of 5-th W.C.E.E., 1973.

[5] Penzien, J. and Watabe, M., "Characteristics of Three-Dimensional Earthquake Ground

Motions," Earthq. Eng. Struct. Dyn. 3, pp. 365-373, 1975.
Ohsaki , Y., Mara , A. and Kiyota, Y., "Stress-Strain Model of Soils for Seismic

Analysis (in Japanese)," proc. of 5-th J.E.E., pp. 679-704, 1978.

Kitagawa, Y. and Matsushima, Y., "Study of Dynamic Ground Characteristics in Range

of Short and Long Periods," Proc. of 7th W.C.E.E., 1980,

66
HURRICANE WIND SPEEDS I.N THE UNITED STATES

by

Martin E. Batts

Larry R. Russell

Emil Simiu

INTRODUCTION

It has been previously been shown that predictions of extreme wind speeds in

hurricane-prone regions cannot in general be based upon the statistical analysis of the

largest annual wind speeds recorded at a given site [11, p. 84]. For this reason, estimates of extreme hurricane wind speeds at a site are commonly obtained by indirect methods

[2,5,12,13] from (1) statistical information on the climatological characteristics of hurricanes, and (2) a physical model of the hurricane wind structure. The climatological

characteristics of hurricanes include: (a) rate of hurricane occurrence in any given

region; (b) difference between atmospheric pressures at the center and at the periphery of

the storm; (c) radius of maximum wind speeds; (d) speed of storm translation (e) direction

of storm motion, and (f) crossing point coordinate along the coast or on a line normal to

the coast. The physical model of the hurricane wind includes assumptions on: (a) dependence of surface wind speeds upon difference between atmospheric pressures at center and

periphery of the storm, radius of maximum wind speeds, speed of translation, latitude, and

position of the point being considered with respect to center of the storm; (b) storm decay

as the storm travels over land and its supply of energy in the form of warm moist air from

the ocean surface is thus cut off; (c) reduction of wind speeds due to friction over land;

and (d) ratios between wind speeds averaged over various time intervals.

Given the available statistical information on the climatological characteristics of

hurricanes, probabilistic models are estimated for each of these characteristics. A large

number of hurricanes is then generated by Monte Carlo simulation (random sampling) from

these probabilistic models, as will be explained subsequently in the paper. The largest

wind speeds occurring within each of these hurricanes at the site of concern represent the

data from which the cumulative distribution function of the hurricane wind speeds at that

67
site is then estimated. This procedure was first developed by Russell [10], who also

developed the computer program used in this work.

The purpose of this paper is to present estimates of hurricane wind speeds on the Gulf

and East Coasts of the United States based on the procedure just outlined and on data taken

from Ref. [4] and [5]. The probabilistic and physical models used to obtain these estimates

will be described, and an analysis will be made of errors inherent in the estimates. Estimates of hurricane wind speeds in California [14] are not treated in this paper.

PROBABILISTIC MODELS

The following probabilistic models were used:

(1) The hurricane occurrence is described by a uniform Poisson process.

(2) The probability distribution of the pressure difference between center and

periphery of storm, APmax > is lognormal. To eliminate values of Apmax judged, in light of

historical data [6], to be unrealistically high, the distribution is censored so that

Apmax < 101.6 mm (4.00 in) of mercury. (Note that Apmax = 101.6 mm corresponds to the lowest atmospheric pressure ever recorded worldwide [15].) The effect of this censoring will

be commented upon subsequently in this paper.

(3) The probability distribution of the radius of maximum wind speeds, R, is lognormal. This distribution is censored so that 8 km < R < 100 km to avoid unrealistically

"tight" or "broad" storms [6]

.

(4) The average correlation coefficient between R and Ap^x is approximately -0.3

(see Ref. [6], pp. 68 and 69). All other climatological characteristics are assumed to be

statistically independent.

(5) The probability distribution of the speed of translation, s, is normal. This

distribution is censored so that 2 km/hr < s < 65 km/hr [6].

(6) The cumulative distribution function of the hurricane crossing point along the

coast is a curve matching the historical data as recorded in Ref. [6]. Separate

distribution curves are defined for entering, exiting, upcoast heading, and downcoast

heading storms. The length of coast being considered for entering and exiting storms

includes a 470 km segment downcoast of the location under investigation, and a 370 km

segment upcoast of that location, the influence of storms not crossing these segments being

68
negligible for practical purposes. The downwind and upwind segments are unequal because,

owing primarily to the presence of the translational speed of the storm, the hurricane

wind structure is asymmetrical. As an example, the distribution curve assumed for entering storms affecting coastal milepost 400 n. mi, (740 km) (see Figure 1) is shown in

Figure 2.

(7) The crossing points on a line normal to the coast are assumed to be uniformly

distributed between 110 km and 280 km from the coast in the case of upcoast heading storms,

and between 110 km and 360 km in the case of downcoast heading storms.

(8) The cumulative distribution functions of heading (direction of storm translation)

for entering storms are curves matching the historical data recorded in Ref . [6]. As an

example, Figure 3 shows the distribution of the storm headings about their mean value for

milepost 400 n. mi. (740 km). For exiting storms, and for storms crossing a line normal to

the coast, the distributions are assumed to be uniform between ± 30° of the mean. The mean

values of the storm headings are obtained from the data of Refs. [4] and [6]. In all cases

it was assumed that the path of the storm motion is a straight line.

PHYSICAL MODELS

Maximum Gradient Wind Speed

The maximum gradient wind speed can be written as:

V = -M + (I dp.)

172

(1) gx 2 p dn *

'

(see Ref. [11], p- 18), where f = Coriolis parameter, dp/dn = pressure gradient, p = air

density, and R = radius of maximum wind speeds. It is reasonable to assume

dp = ctAp (2)

dn max

where a is a coefficient determined empirically. It then follows from Equations 1 and 2

V = K Ap - M (3)

gx max 2

where the notation K = [R a/p]^'^ is used. Values of K consistent with those determined

empirically by the Hydrometeorological Branch (National Weather Service, National Oceanic

69
and Atmospheric Administration) and used in Ref . [9] vary between 6.97 m/s/mm1 / 2 (77.5 mph/

in1 / 2 ) at latitude 23°N to 6.93 m/s/mm 1 / 2 (77.0 mph/in L / 2

) at latitude 45°N. [These values

may be used in conjunction with the assumption that the pressure at the periphery of the

storm is pn

= 765 mm (29.77 in).]

Wind Speeds at 10 m Above the Ocean Surface

The maximum wind speed at 10 m above the ocean surface, averaged over 10-min, is

assumed to be given by the empirical relation [9]

V(z = 10, R) = 0.865 V gx + 0.5 s (4)

Let the center of the storm be denoted by 0, and let a line OM be defined that makes

an angle of 115° clockwise from the direction of motion of the storm. The 10-min wind

speed at 10 m above the ocean surface at a distance r from 0 along the line OM is denoted

by V(z = 10, r). The ratio V(z = 10,r)/V(z = 10, R), is assumed to depend on r as shown in

Figure 4 [9]. Let now the angle between a line ON and line OM be denoted by 6. The 10-min

wind speed V(z = 10, r, 6) at 10 m above the surface at a distance r from the storm center on

line ON is a assumed to be given by the expression [9]

V(z = 10, r, 8) = V(z = 10, r) - s (1 - cos 9) (5)

2

The wind velocity vector has a component directed toward the center of the storm. The

angle between that vector and the tangent to the circle centered at 0 is assumed to be

between 0° and 10° in the region 0 < r < R, between 10° to 25° in the region R <^ r < 1.2 R,

and 25° in the region r << 1.2 R [9].

Storm Decay

The storm decay was assumed to result from a decrease with time of the difference

between pressure at the center and at the periphery of the storm, given by the following

relation

Ap(t) = A Pmax - 0.02 [1 + sin $]t (6)

where t = travel time in hours, Ap(t) and Apmax are given in inches, and § = angle between

coast and storm track (0 < <p < 180°). As shown subsequently in this paper, results based

70
on this model were compared with results based on Malkin's model, which is consistent with

measurements reported in Ref. [8]. This comparison appears to support validity of Eq. (6)

as a conservative approximation.

Reduction of Wind Speeds Due to Friction Over Land

The factor for reducing overwater to overland surface wind speeds used in Ref. [9] is

0.78 for wind speeds over 37 m/s. In Ref. [3], the ratio V

£

( 10)/VW( 10) , where V(10) =

10-min mean speed at 10 m above the surface, and the superscripts w and I indicate "overwater" and "overland", respectively, is given as

where the retardation factor p = 0.83 and the roughness length z 0 = 0.005 m, so that

V^( 10)/VW( 10) = 0.79, i.e., this ratio has almost exactly the same value as in Ref. [9].

Note that Eq. (7) is based on similarity relations applicable to storms in which the

gradient wind may be regarded as geostrophic, so that its applicability to hurricane

winds is uncertain, at least in the region of maximum wind speeds. In this work it was

therefore assumed conservatively that V^( 10)/VW( 10) = 0.85.

Dependence of Wind Speeds Upon Averaging Time

Research results concerning the ratios of hurricane wind speeds corresponding to

different averaging times do not appear to be available. In the absence of such results,

it was judged reasonable to use wind speed ratios obtained for non-tropical storms by

Durst which are summarirzed, e.g., in Ref. [11], p. 62. For example, the ratio between

the fastest one-minute and the fastest 10-minute speed at 10 m above open terrain was

assumed to be 1.18.

To estimate probabilities of occurrence of hurricane wind speeds, one thousand

hurricanes were assumed to hit the area adjoining each site being investigated. (Recall

that this area was defined previously in this section "Probabilistic Models," item 6 and

7.) Four categories of storms were considered: entering hurricanes (i.e., hurricanes

V

£ (10)

Vw(10) 0.2 p In iP_

z Q

1 (7)

PROBABILITIES OF OCCURRENCE OF HURRICANE WIND SPEEDS

71
moving inland), exiting hurricanes (i.e., hurricanes headed toward the ocean after moving

over land), upcoast heading hurricanes (i.e., hurricanes whose center remains offshore and

moving upcoast), and downcoast heading hurricanes. For each location, the ratios between

the number of hurricanes belonging to each of these categories and the total number of hurricanes affecting that location were estimated from the information given in Ref . [4] and

[6].

The climatological data for each of the one thousand hurricanes were determined from

the respective probability distributions by Monte Carlo simulation (random sampling).

Associated with each hurricane is a wind field, and a set of wind speeds at the site of

interest which depend upon the position of the moving hurricane center with respect to the

site. These wind speeds were calculated for a sufficiently large number of such positions

(6 or 7, depending on track). The largest among these speeds represents the maximum wind

speed caused by the hurricane at the site. A set of one thousand such wind speeds is thus

obtained, which is used as the basic set of data for the estimation of the probability of

occurrence of hurricane wind speeds at the site. (This procedure was applied to estimate

extreme wind speeds without regard to direction, as well as extreme wind speeds blowing in

specified directions.) The wind speeds are ranked by magnitude, and the probability of

occurrence of the i-th wind speed v^, in a set of m wind speeds (m = 1,000 in the work

reported herein) is obtained as follows.

Let the probability that the wind speed in any one storm is less than v be denoted

by F v

. The probability that the highest wind speed V in n storms is less than v can be

written as

P(V < v|n) = F* (8)

Let the probability that V < v in x years be denoted by P(V < v, x). It is possible to

write

P(V < v, t) =

I P(V < v|n) P(n, t) (9)

n=0

where P(n,x) denotes the probability that n storms will occur in t years. If a Poisson

process is assumed to described P(n,x), and if use is made of Eq . (9),

72
n (vx) n e"VT P(V < v,i) = I F$

n=0 n!

-VT (vtF

v

)' n=0 n!

= e -VT e VTFv

= e- VT ( 1 - F v) (10)

where v = annual rate of occurrence of hurricanes in the area of interest for the site

being considered. For t =1, P(V < v,x) = probability of occurrence of wind speeds less

than v in any one year.

Consider now the wind speed . Its probability of occurrence can be written as

(11) m + 1

Thus

i " v (l - m + l)

P(V < Vi> l) = e (12)

The estimated mean recurrence interval of the speed is then

N = I (13)

i " m + l)

1 - e

NUMERICAL RESULTS

Extreme Wind Speeds

Hurricane fastest-mile wind speeds at 10 m above ground over open terrain were

estimated for 58 mileposts (from 150 through 3000 in increments of 50 - see Figure 1) at

the coastline, and at 25 km, 50 km, 100 km and 200 km inland. The results of the calculations were smoothed as follows. For each milepost the average was calculated of the

fastest-mile wind speeds at that milepost and at its two neighboring mileposts [located

each at 50 n.mi. (83 km) from it]. This averaging was done, for each milepost, at the

73
coastline, as well as at 25 km, 50 km, 100 km, and 200 km inland. A least-squares straight

line was fitted at each milepost to the five averages thus obtained. The ordinates of

this line at the coastline and at 200 km inland represent the values given by the curves

of Figures 5 and 6, respectively. Extreme fastest-mile speeds at distances inland of less

than 200 km from the coastline may be obtained by linear interpolation between Figures 5

and 6. It is noted that the values of Figures 5 and 6 differ by at most a few percent

from the corresponding unsmoothed values originally calculated for the coastline locations.

The results shown in Figures 5 and 6 represent smoothed estimates of hurricane wind

speeds without regard to direction. As previously indicated, similar results were obtained

for winds blowing in specified directions. As an example, estimates of coastline fastestmile hurricane wind speeds at 10 m above ground over open terrain at milepost 600, smoothed

in the same manner as the results of Figures 5 and 6, are plotted in Figure 7 as a function

of direction. Similar plots for coastline locations at mileposts 150 through 2900 (in

increments of 50) are included in Ref. [1].

i Errors in the estimation of extreme hurricane wind speeds are dealt with in the

subsequent section. It is noted here that these errors increase as the mean recurrence

interval increases. Therefore, estimates of wind speeds corresponding to mean recurrence

intervals exceeding 100-yr, say, may contain significant errors. Nevertheless, such estimates obtained by the rational procedure used in this paper are of interest as they provide

the only available basis for assessing the validity of wind load factors currently being

used for structural design purposes in hurricane-prone regions.

Probability Distributions of Hurricane Wind Speeds

The numerical results obtained from the calculations were used to estimate probability

distributions of hurricane wind speeds (blowing from any direction). It was found, by

using the probability plot correlation coefficient method, that in all cases the best

fitting distribution for the 1-minute speeds was the Weibull distribution with tail length

parameter y > 4 (for practical purposes y = 4 can be assumed, the differences between

results based on y = 4 and y > 4 being negligible). As an illustration, the results

obtained at milepost 450 [1-minute wind speeds at coastline over water, in knots (1 knot =

1.9 km/hr)] are plotted on Weibull (= 4) and Extreme Value Type I probability paper on

Figures 8 and 9, respectively. It is seen that the results fit the Extreme Value Type I

74
distribution poorly, i.e., the results do not fit a straight line on Figure 9, as they do

on Figure 8.

MIXED PROBABILITY DISTRIBUTIONS OF HURRICANE AND NON-HURRICANE WIND SPEEDS

Mixed probability distributions of hurricane and non-hurricane fastest-mile (as

opposed to 1-minute) wind speeds were estimated by using the following expression:

P(V < v) = P H (VH < v) • PnhCVnh < v) (14)

in which Py (Vjj < v) = cumulative distribution of hurricane wind speeds Vjj, and

P NH ( VNH < v) = cumulative distribution of non-hurricane wind speeds, V^. The distribution Pji was determined as shown in this report. For the distribution P^jh, an Extreme

Type I distribution was assumed. The parameters of P^ were estimated at various locations from maximum annual wind speed data that did not include hurricane wind speeds. The

results of the calculations showed that the effect of the non-hurricane winds is negligible

for mean recurrence intervals of the order of 50 years or more (see Figure 10). For mean

recurrence intervals of about 20 years, the estimated wind speeds that include the effect

of non-hurricane winds exceed the estimated hurricane speed by about 5 percent. Note that

these conclusions are not applicable north of Cape Hatteras, where non-hurricane wind may

control the design at certain locations.

ESTIMATION ERRORS

The errors in the estimation of hurricane wind speeds by the procedure used in this

work may be divided into four categories: (1) sampling errors, Que to (a) the limited size

of the data sample used in making statistical inferences on the climatolcgical characteristics of hurricanes (in the United States these data samples correspond to lengths of

record of 75 to 100 years), and (b) the limited number of hurricanes generated by the

Monte Carlo simulation; (2) probabilistic modeling errors , due to the imperfect choice of

the distribution functions to which the climatology data are fitted (e.g., assuming that a

lognormal distribution holds when in fact a normal distribution would be more appropriate);

(3) observation errors, due to the imperfect measurement or recording of the true values of

the climatological characteristics; and (4) physical modeling errors , due to the imperfect

75
representation of the dependence of the wind speed upon the various climatological

characteristics and micrometeorological parameters.

Sampling Errors

As shown in some detail in Ref . [2], the standard deviation of the errors associated

with the limited size of the climatological data samples (corresponding to about 75 to 100

years of record) is of the order of 6 percent to 10 percent of the estimated wind speeds,

while the standard deviation of the errors due to the limited number, m of hurricanes used

in the simulation is of the order of a few percent for m ~ 1,000. In some cases, say, at

Jacksonville, Florida, the poorly understood and seemingly anomalous occurrence histories

may lead to larger errors than are typical of the model overall. Part of the sampling

errors due to the limited number of hurricanes used in the simulation is eliminated by the

use of the smoothing procedure described in the preceding section.

Probabilistic Modeling Errors

To assess the effect of various probabilistic models used in the simulation,

calculations were carried out at mileposts 400, 1450 and 2100 (see Figure 1) under the

following assumptions:

(a) Distribution of Apmax is normal. This assumption results in a decrease of the

50-yr, 100-yr, and 400-yr wind speeds with respect to the corresponding values based on the

assumption of lognormality by about 5 percent, 5 percent, and 8 percent, respectively.

(b) Distribution of R is normal. This assumption results in an increase of the

50-yr, 100-yr, and 400-yr wind speeds with respect to their corresponding values based on

the assumption of lognormality by about 1 percent, 2 percent, and 6 percent, respectively.

(c) Distribution of Apmax and R are both normal. This assumption results in a

decrease of the 50-yr, 100-yr, and 400-yr wind speeds with respect to their corresponding

values based on the assumption that Apmax and R are both lognormally distributed by about

3 percent, 3 percent, and 4 percent, respectively.

(d) Distribution of s is lognormal . The differences between estimates of wind speeds

based on this assumption on the one hand, and on the assumption that s is normally

distributed on the other hand, are negligible for practical purposes.

76
(e) Distribution of Apmax is not censored. The censoring of the distribution of

A Pmax ( A Pniax < 101.6 mm) reflects the writers' belief that this upper bound is dictated by

physical considerations, and that an unbounded distribution would thus be physically

unrealistic. Nevertheless, calculations were carried out for five stations in which it

was assumed that the lognormal distribution of Apmax is unbounded. It was found that the

hurricane wind speeds and their cumulative distribution functions were for practical

purposes the same regardless of whether the condition Apmax < 101.6 mm was assumed or

not

.

A final observation with regard to the choice of probabilistic models concerns the

representation of the hurricane occurrences as a Poisson process. It can be shown that

this model could be replaced, e.g., by a geometric probability model, without any significant effect upon the estimated results, particularly for winds with mean recurrence

intervals of the order of 20 years or more.

Observation Errors

To assess the effect of possible errors in the measurement (or recording of the

parameters Apmax and R, calculations were carried out at mileposts 400, 1450, and 2100

(Figure 1) in which it was assumed that the mean and/or standard deviation of Apmax and R

are larger than those estimated from the data of Ref. [6]. Increasing both the mean and

the standard deviation of Apmax by a factor of 1.1 resulted in an increase in the estimated values of the 50-yr, 100-yr, and 400-yr winds of about 5 percent at milepost 400,

8 percent at milepost 1450, and 10 percent at milepost 2100. Leaving the mean unchanged

and increasing the standard deviation by a factor of 1.1 resulted in an increase of the

estimated speeds of the order of a few percent. The effect of increasing or decreasing the

mean and/or the standard deviation of the radii of maximum wind speeds, R, by a factor as

high as 1.5 was of the order of a few percent at most.

As previously indicated, the data used in this report were obtained from Ref. [6].

The estimation of hurricane wind speeds at various locations could presumably be improved

by using additional sources of climatological information, e.g., those available in the

archives of the Corps of Engineers. The writers believe that this is the case particularly

for South Texas, Louisiana, southwest Florida and Jacksonville.

77
Note that estimated hurricane wind speeds are lower at inileposts 1000 and 1750

(Figures 1, 5, and 6) than in the regions adjacent to these mileposts. This may be due to

the relation between configuration of the coast and the hurricane paths. However, it may

be that hurricanes at and near these mileposts were underreported in the past. It appears

therefore prudent to increase the wind speeds calculated at mileposts 1000 and 1750 by

about 5 mph or 10 mph. Finally, it was found that the storm decay overland was very weak

at mileposts 650 and 2250. This is due, at least in part, to the configuration of the

coast at these mileposts, although the spurious occurrence of strong simulated storms may

also have contributed to this result.

Physical Modeling Errors

The coefficient K in Eq. (3), the coefficient of Eq. (4), and the dependence of wind

speeds upon radius represented in Figure 4, have been determined empirically by the

National Weather Service on the basis of careful correlations of pressure and wind speed

measurements. Clearly, the corresponding physical model of the hurricane wind field is,

nevertheless, imperfect. It is the writers' belief, based, e.g., on information from

Ref. [7], that the standard deviation of the errors inherent in this model is of the

order of 5 percent to 10 percent.

It was noted previously that the coefficient K assumed in this work may be used in

conjunction with the value p n

= 756 mm (29.77 in). To check the effect on the calculated

results of the assumption p n

= 760 mm (29.92 in), calculations based on this assumption

were carried out at mileposts 400, 1450, and 2100. It was found that differences between

results based on p n

= 756 mm and p n= 760 mm were of the order of 3 percent.

It was noted that results based on the model for storm decay used in this work

(Eq. (6)) were compared with results based on Malkin's model, which is consistent with

measurments reported in Ref. [8]. The differences between the two sets of results were in

all cases small (of the order of ± 2 percent).

In estimating wind speeds over water immediately off the shoreline it was assumed that

the roughness length parameters corresponds to "flow over water" conditions in all

directions (this will be referred to as "surface friction hypothesis I"). In reality,

each location immediately offshore is affected not only by hurricane winds which blow from

the ocean (and for which the assumption that a roughness length corresponding to "flow
over open terrain" conditions may have to be used. The hypothesis that the roughness

length depends upon whether the winds flow from the ocean or from land will be referred to

as "surface friction hypothesis II." whether or not "surface friction hypothesis II" is

warranted is by no means always certain: indeed, near the shoreline the terrain is likely

to be flooded during the occurrence of hurricanes. Nevertheless, it is of interest to

check the differences between results based on surface friction hypotheses I and II.

Calculations carried out at mileposts 400, 850, 1450, 2100, and 2800 showed that estimated

wind speeds corresponding to hypothesis II were lower than those based on hypothesis I by

an amount generally not exceeding 3 percent.

Local wind intensification due to the presence of rainbands occur at all locations

south of the 30° north parallel. These intensifications might have to be accounted for by

increasing the wind speeds calculated herein by as much as 10 mph (4.47 m/s) or so.

As previously noted, the physical models used in this report, and particulalry the

storm decay model, provide a description of the wind field that is probably less accurate

at sites located north of Cape Hatteras, N.C., (mileposts 2200 or more, see Figure 1). The

numerical results obtained for these sites should therefore be viewed very cautiously.

The physical model of the hurricane wind speeds may be refined by using a detailed

representation of the coastline and its adjacent bodies of water. Such refinements may

result in an improvement of hurricane wind speed estimate by an amount of the order of

5 percent or so.

Additional refinements might include accounting for slower storm decay where large

bodies of water are present inland, e.g., in the low-lying areas of Louisiana. Additional

research is required to develop improved storm decay models in such situations.

CONCLUSIONS

In this paper, estimates are presented of hurricane wind speeds along the Gulf and

East Coasts of the United States. The paper describes the sources of data, the probabilistic models for the climatological characteristics of hurricanes and the physical models

for the hurricane wind speed field used in the estimations. Estimated values of fastestmile hurricane wind speeds at 10 m above ground in open terrain at the coastline and at

200 km inland are given for various mean recurrence intervals (Figures 1, 5, and 6).

79
Fastest-mile wind speeds at distances inland of less than 200 km from the coastline may be

obtained by linear interpolation between these values. It was found that the estimated

hurricane wind speed are best fit by Weibull distributions with tail length parameters

Y ^> 4, rather than, e.g., by Extreme Value Type I distributions. Estimates are given of

various errors inherent in the estimated values of the hurricane wind speeds. The confidence bands for the estimates were found to be of the order of at least ± 10 percent at

the 68 percent confidence level. It was also noted that, owing to the possible inapplicability of the physical models used in this work at locations north of Cape Hatteras,

estimated hurricane wind speeds given for these locations should be viewed with considerable caution. In certain special cases it may be desirable to base estimates of hurricane wind speeds on more elaborate data, physical models, and representations of the

coastline than those used in this work. Nevertheless, in the writers' opinion, the results

presented in this paper provide a rational and consistent basis for making realistic decisions on the specification of wind speeds and load factors in hurricane-prone regions.

ACKNOWLEDGMENTS

This material is based upon activities supported by the National Science Foundation

under Agreement No. ENV-7716113 and by the Department of Energy, Office of Assistant Secretary, Conservation and Solar Applications. Any opinion, findings, and conclusions or

recommendations expressed in this publication are those of the authors and do not

necessarily reflect the views of the National Science Foundation or of the Department of

Energy. The writers wish to acknowledge the collaboration of M. R. Cordes and J. R. Shaver

of the National Bureau of Standards, who have contributed to the planning and execution of

the computations.

REFERENCES

[1] Batts, M. E. et al .

, Hurricane Wind Speeds in the Gulf and East Coasts of the United

States , Building Science Series Report, National Bureau of Standards, Washington,

D.C., 1980.

[2] Batts, M. E., Cordes, M. R. , and Simiu, E., "Sampling Errors in Estimation of Extreme

Hurricane Winds," Proc. of the 3rd Engineering Mechanics Division Specialty

Conference, September 17-19, 1979, Austin, Texas.

80
[3] Bietry, J. Sacre , C, and Simiu, E., "Mean Wind Profiles and Change of Terrain

Roughness," Journal of the Structural Division , ASCE, Vol. 104, No. ST10, October

1978, pp. 1585-1595.

[4] Cry, G. W. , Tropical Cyclones of the North Atlantic Ocean-Tracks and Frequencies of

Hurricanes and Tropical Storms , 1871-1963, Technical Paper No. 55, U.S. Department of

Commerce, Weather Bureau, Washington, D.C., 1965.

[5] Gomes, L., and Vickery, B. J., On the Prediction of Tropical Cyclone Gust Speeds

Along the Northern Australian Coast , Research Report R278, School of Civil

Engineering, University of Sydney, February, 1976.

[6] Ho, F. P., Schwerdt, R. W., and Goodyear, H. V., Some Climatological Characterstics

of Hurricanes and Tropical Storms, Gulf and East Coasts of the United States , NOAA

Technical Report NWS 15 , Washington, D.C. May 1975.

[7] Holliday, C, On the Maximum Sustained Winds Occurring in Atlantic Hurricanes , Technical Memorandum No. WGTM-SR-45, U.S. Department of Commerce, Environmental

Sciences Services Administration, Weather Bureau, Fort Worth, Texas, 1969.

[8] Malkin, W. , Filling and Intensity Changes in Hurricanes on Land , National Hurricane

Research Project, Report No. 34, U.S. Department of Commerce, Washington, D.C,

November 1959.

[9] Revised Standard Project Hurricane Criteria for the Atlantic and Gulf Coasts of the

United States , Memorandum HUR7-120, U.S. Department of Commerce, NOAA, June 1972.

[10] Russell, L. R., "Probability Distributions for Hurricane Effects," Journal of the

Waterways, Harbors, and Coastal Engineering Division , ASCE, Vol. 97, No. WW1,

February 1971, pp. 139-154.

[11] Simiu, E., and Scanlan, R. H., Wind Effects on Structures , Wiley-Interscience

,

New York, N.Y. , 1978.

[12] Sklarin, J., Stochastic Analysis of Hurricane Wind Loads (Harman, T. G., and

Cornell, C. A., supervisors), Report R77-2, Order No. 563, Department of Materials

Science and Engineering, Massachusetts Institute of Technology, Cambridge,

Massachusetts, January, 1979.

[13] Tryggvason, B. V., Surry, D., and Davenport, A. G., "Predicting Wind-Induced Reponse

in Hurricane Zones," Journal of the Structural Division , ASCE, Vol. 102, No. ST12,

December 1976, pp. 2333-2350.

[14] Windstorms in California, State of California, Department of Water Resources,

Sacramento, California, December, 1979.

[15] Worldwide Extremes of Temperature, Precipitation and Pressure Recorded by Continental

Area , ESSA/P1680032 , Environmental Data Service, U.S. Department of Commerce

October, 1968.

APPENDIX I - NOTATION

dp/dn = pressure gradient

f = Coriolis parameter

K = coefficient

p = retardation factor

81
pn

= standard pressure

P(n,x) = probability that n storms occur in t years

P(V < v|n) = probability that the wind speed V in no storms is less than v

P(V < v/t) = probability that the wind speed V in t years is less than v

r = distance from storm center

R = radius of maximum wind speeds

s = hurricane translational speed

t = time, in hours

VgX

= maximum gradient windspeed

V(z,R) = maximum velocity at z meters above ground at the radius of maximum winds

V(z,r) = maximum velocity at z meters above ground at the radius r from the eye

V(z,r,6) = velocity at z meters above ground at radius, r and angle 9 from OM

z = height above ground or water surface, in meters

a = coefficient

Apmax = pressure drop from periphery of hurricane to its center

p = air density

0 = angle between lines ON and OM

<j> = angle between storm track and coastline

£ = angle between storm track and coastline

82
Figure 1. Locator map with coastal distance intervals marked

in nautical miles (1 n. mi. =1.9 km) [5].

83
100 200 300 400 500 600

C0RST X-INO POINT (N.M.

)

Figure 2. Cumulative distribution function of entering hurricane crossing

point, milepost 400 n. mi. (1 n. mi. « 1.9 km).

-100 -50 0 50 100

HEADING ABOUT HERN fDEC)

gure 3. Cumulative distribution function of entering storm headings about

mean, milepost 400 n. mi. (1 n. mi. - 1.9).

84
0.1 0.2 0.3 0.4 0 5 0.6

V|10,r)

V(IO.R)

0.7 0.8 0.9 1.0

Figure 4. Ratios V(10 ,r) /V(10 ,R)

nPH is*-

125

?5

sr

25-

2000 VR

- 100 VR 50 VR

25 VR

10 VR

0

l ii i i

|

i ii i

| iii i| 'i'i|'i'i|i i ii|ii'»;<»ii]i'''|'iii)'ii»|'ii')

350 500 750 1000 1250 1500 1750 2000 2250 2500 2750 30M

HIIEPOST

Figure 5. Estimated fastest-mile hurricane wind speeds blowing from any

direction at 10 m above ground in open terrain near the coastline, for various mean recurrence intervals (1 mph - 0.72 m/s)

.

85
uph

125

1*0

75

SO

25-

\ / V

see* y*

tee vr se yr as VR

te vr

r

'

I

1 1 1 ' I '

1 1 1

1 1 '

1 ' I

1 '

1 ' i1 '

1 1 i

1 '

1

1

i

1 1 » i 1 1 ' 1

I

1 ' ' '

i

1 1 1 1

I 1 1 ' '

i 2se see 75* teee i25e isee !7se 2eee 22se 2see 27se w

HILEPOST

Figure 6. Estimated fastest-mile hurricane winds speeds blowing from any direction

at 10 m above ground in open terrain at 200 km inland, for various mean

recurrence intervals (1 mph * 0.72 m/s) .

MPH

80

60

40

20

100 Year

10 Year

j i I I I 80 100 270 300

Degrees from North (Clockwise)

Figure 7. Estimated fastest-mile hurricane wind speeds blowing from various

directions at 10 m above ground in open terrain near the coastline,

for various mean recurrence intervals, milepost 600 n. mi. (1 n. mi,

- 1.9 km; 1 mph * 0.72 m/s)

Ho
REDUCED (MR t ATE 1.25

1.20

UEIBULL PROBABILITY PLOT RETURN PERIOD

1.15

1.10

1.0S

1.00

0.95

0.90

.999886

.899998 .998989

.995968

. WWW

.988699

.958999

.988998

5888

2989

1988 588

298

199

20. 40. 60. 80. 100. UINDSPEED (1 HIN. ) KNOTS

Figure 8. Weibull probability distribution plot of 1-min hurricane wind speeds,

mi lepost 450 n . tni .

REDUCED UARIATE FREQUENCY EXTREME TVPE I PROBABILITY PLOT RETURN PERIOD

8.00

7.00

e.00

5.00

4.00

3.00

2.00

1.00

0.00

-1.00

10. 30. 50. 70. 90. UINDSPEED (1 WIN. ) KNOTS

- .999SM

- .99958* I

- mtnn

- .998899

- .995898

- .999849

- .5SBWJB

- ,98»»»8

- .89*9*9

- .56*898

110.

3M

LM

59

21

18

5

Figure 9. Extreme value Type I probability distribution plot of 1-min hurricane

wind speeds, milepost 450 n. mi.

87
REDUCED UAR1ATE FREQUENCY EXTREME TVPE I PROBABILITY PLOT

7.00

6.00

5 .0®

4 . Mt.

3.00

3.00

1.00

0.00

-1.00

- .998989

- .999999

- .988898

.999599

.995998

- . 999999

,41.

-C0HB1 lEU

-HnNHIIWTfflHT

ANE

RETURN PERIOD

eeae

1999

598

40. 45. 50. 55. 60. 65. 70. 75. 80.

VELOCITY

COMBINED HURRICANE AND NONHURRICANE PROBABILITY

299

189

S®

26

18

5

Figure 10. Extreme value Type 1 probability diatribution of faateat-mile wind

apeed a

.

88
A STATISTICAL RESULT ON MAXIMUM SURFACE WINDS SURROUNDING TYPHOONS

by

Shigemi Fujiwhara

I. Subbarramayya

ABSTRACT

Reconnaissance aircraft data from tropical storms and typhoons in the western North

Pacific for the period 1974-78 were examined. Regression equations for maximum surface

winds, using central pressure and the squareroot of the central pressure-depth as

independent variables, were derived and compared.

89
INTRODUCTION

The strong wind in a tropical cyclone is the main factor that is responsible for the

extensive damage caused by cyclones. The strong winds are also responsible for strongsurges which cause coastal inundations and loss of life. Therefore, accurate estimations

of maximum surface winds associated with cyclones is an important problem to the tropical

storm forecaster.

Horiguti (1926) suggested an empirical equation for the radial pressure profile of

typhoons. Using that equation and the cyclostrophic wind assumption Takahashi (1948)

showed that

,

where, Vmis the maximum surface wind in m/ sec , pn

is the peripheral pressure in mbs and

p 0 is the central pressure. But, earlier Takahashi (1940) obtained 5.1 for the constant

on the right hand side of Eq. (1) from statistical study of observations. However, he

suggested in 1948 that a value of 6 for the constant Is safer. From similiar studies on

hurricanes in Atlantic, Fletcher (1955) obtained the equation,

where, V m is in kts and p n and p 0 are as in Eq. (1).

The above equations were based on very few observations in the region of strongest

winds. They are primarily dependent on an extrapolation of the conditions around areas of

the strongest winds.

With the advent of reconnaissance flights into tropical cyclones in the mid-1950 1 s

by the U.S. Air Force, a larger number of observations of central pressures and maximum

surface winds have become available. Central pressure is obtained either directly by

drop-sonde, or, by extrapolation from the temperature and height data at a 700 mb (flight)

level. The surface winds, on the other hand, are estimated by comparing the state of the

sea-surface with standard photographs. The individual maximum wind reports, therefore,

are less accurate (Shimada, 1963) and contain subjective errors. However, reconnaisance

observations are the best data available at present on maximum surface winds in tropical

cylcones

.

Vm = 4.5 pn

- p 0 (1)

Vm

= 16 / p n

- p 0 (2)

90
Shea and Gray (1973) examined maximum winds in relation to central pressures using

aircraft observations and indicated an inverse relationship between the two.

The authors in this paper have therefore made a careful study of the relationship

between V mand p 0 using reconnaissance observations and obtained an empirical equation

relating the two. There are factors other than central pressure which affect the maximum

surface wind (Riehl, 1963), but those factors are of secondary importance. The authors

have deliberately ignored the other factors in the present study, because, data on such

parameters are less available and the future applicability of the envisaged equation would

be restricted.

DATA AND ANALYSIS

The reconnaissance data for all the storms and typhoons of the western area of the

North Pacific for the years 1974-78 during the typhoon season (July-November), available

at the Japan Meteorological Agency, have been examined. The total number of useful observations, i.e., simultaneous central pressure and maximum surface wind, were about 510 and

these are distributed approximately over the area between 10°-30°N and 110°-170°E. Though

several flights may be made into a single cylcone by the Joint Typhoon Warning Center,

Guam, not all are successful and a full series of observations could not be made in all the

successful penetrations.

A scatter-diagram between maximum wind and central pressure was first prepared

(Figure 1). The scatter is large, but the observations show a regular trend. To ascertain

the nature of the trend, the total pressure range was divided into 10 mb intervals and the

mean pressure and mean wind in each interval were calculated These mean-value points show

(Figure 2) a generally linear trend. Therefore, regression equations for Vm

with / 1010-p o

and p 0 as independent variables have been evaluated and compared.

RESULTS AND DISCUSSION

The number of observations and the standard deviation for wind speeds in each pressure

interval are presented in Table 1 . The number of observations per class interval is more

than 100 in weaker storms; this decreased to 5 in the extremely severe typhoon. The standard deviation of winds in each class generally increased along with the wind speed, but in

91
the very strong wind region, it is less than the deviation at relatively lower winds.

This may be because the reconnaissance flights are less successful in very severe typhoons.

The estimated winds using Fletcher's equation have been found to be quite high

compared to the observed winds. This is in conformity with the reports of other workers

(Dunn and Miller, 1960). Winds estimated by the formula recommended by Takahashi show

better agreement with observations, but the standard deviation of error estimates is

found to be 17.87 kts.

The best-fit linear regression between V m and / (1010 - p Q

) for the original

observations is

where, Vm

in knots and p0 is in millibars. The standard deviation of errors of estimates

of this equation is 15.87 kts. The best-fit linear equation for Vm

with p 0is

The standard deviation of errors of estimates of this equation is 15.85 kts, which is

almost the same as for Eq. (3). Thus there is no statistically significant difference

between the two equations, but it may be noted from Figure 2, that the mean-value points

are randomly situated on either side of the straight line, while they are consistently

below the curved line in the medium-wind range and above in the high-wind range. This

means, Eq. (3) overestimates the intensity in the case of moderate typhoons and

underestimates in the case of severe typhoons.

The weaker observed winds in very deep typhoons compared to those predicted by Eq. (4)

could be partly due to the weakening of the autumn typhoons that have moved over to the

Philippines and Taiwan land areas while the pressure depth is maintained by upper divergence, partly due to the recurvature of the autumn typhoons at relatively low latitudes

and the subsequent structural changes in the typhoons, and partly due to the possible bias

of the sampling of reconnaissance observations towards lower speeds.

It may be remembered that in either Takahashi's or Fletcher's study a linear

relationship between Vm and / (pn

- p 0 had resulted only because of the cyclostrophic wind

assumption they made to obtain an equation for the maximum wind from the empirical pressure

Vm

= 12.4 / (1010 - p Q ) - 8.8 (3)

Vm

= (1010 - Po ) + 25 (4)

92
profiles. Though it appears simple reasoning that V , which is a measure of kinetic

energy, may be directly proportional to the pressure depth, there are other factors like

frictional force that play an important role. The actual mechanics of cyclones are quite

complex and are not yet properly understood. The linear relationship between Vm

and pQ ,

indicated above, points out the complexity of cyclone dynamics. On the other hand, it is

simple to apply in forecasting operations, and is quite useful in typhoon intensity studies

since p 0 can be directly replaced by V m

and vice versa.

When we checked the sea surface wind data over typhoon areas which was obtained by the

microwave scatterometers of Seasat, the 1978 U.S.A. experimental oceanic satellite, we

found this data coincided more closely to the ship observations, but island or land observations were observed to be lower than the ship and the Seasat data because of orographic

deformations

.

We also found that a suitable modification of the deformations produced greater

accuracy with island and land observations in comparison to sea surface winds. Observations of Takahashi's and Fletcher's papers, mentioned above, were almost all island and

land data, and this kind of modification was not taken into consideration at that time.

REFERENCES

Dunn, G. E. and B. I. Miller, 1960, "Atlantic Hurricanes," Louisiana State University

Press, p. 1959.

Fletcher, R. D., 1955, "Computation of Maximum Surface Winds in Hurricanes." Bull.

Amer. Met. Soc . , 36, 247-250.

Horiguti, Y., 1926, "On the Typhoons of the Far East." Mem. Imp. Mar. Obs . , Kobe, 2, No. 3, 111-162.

Riehl, H. , 1963, "Some Relationships between Wind and Thermal Structure of Steady

State Hurricanes." J. Atmos. Sci., 20, 276-287.

Shea, D. J. and W. M. Gray, 1973, "The Hurricane's Inner Core Region. I. Symmetric

and Asymmetric Structure." J. Atmos. Sci., 30, 1544-1564.

Shimada, K., 1963, "Some Problems in the Use of Data Obtained by Aircraft Typhoon

Reconnaissance Flights." Proceedings of the Inter-Regional Seminar on Tropical

Cyclones, Tokyo, Jap. Met. Agency, 307-315.

Takahashi, K., 1940, "Distribution of Various Meteorological Elements in a Typhoon."

Jour. Met. Soc. Japan, 18, 125-130.

1948, "Typhoon in Japan." The Geophy. Mag., 17, No. 1-2,

1-35.

'

93
Table 1

Class 1000 1000- 990- 980- 970- 960- 950- 940- 930- 920- 910- goo- Interval 990 980 970 960 950 4411 930 920 910 900 sgo

No. of Obs. 47 127 115 84 39 19 31 17 8 13 7 5 S.D.

8.2 11.7 15.5 17.3 17.2 24.2 19.8 23.9 20.5 12.8 19.9 19.9

Fig. 1. Central Pressure versus Maximum Surface Wind (Scatter Diagram)

140

120-

100-

80

6C

40-

20

»> «• • • • • • •

. iiy. i ».» • .

.»«. .ni.t .

.

.1»:':.:Hi;;a.

• j

95o 920 940 960 980

CENTRAL PRESSURE IN MBS

1000

94
Fig. 2. Central Pressure versus Maximum Surface Wind (Statistical Curves)

9*00 9*20 9'40 9'60 9*80 lO'OO

CENTRAL PRESSURE IN MBS

V = (1010 - p„) + 25 m 0

V = 12.4/ (1010 - p~) -8.8

m u

95
SURFACE WINDS FROM HURRICANE FREDERIC:

AN ENGINEERING VIEWPOINT

by

Timothy A. Reinhold

ABSTRACT

Surface wind speed data taken during the passage of Hurricane Frederic were obtained

from ten different sites including four sites near Mobile, Alabama and six sites along the

coasts of Mississippi, Alabama and Florida. The form of the recording varies from site to

site as does the exposure, elevation and type of anemometer used. Most sites were visited

in order to estimate the type of terrain, to determine the elevation and type of

anemometer, and to note the location of any potential obstructions.

Each record is analyzed to determine peak to mean ratios between maximum gusts and 20

to 30 minute average wind speeds. These ratios compare favorably with Durst 's relationships between wind speeds for different averaging times using the corrections for different

elevations and types of terrain given by Simiu and Scanlan. The wind data from each site

was subsequently corrected to standard conditions, i.e. open terrain exposure at 10 meters

elevation, using the site characterization and Durst's relationships as modiifed by Simiu

and Scanlan. Consistency of different types of wind speed data is illustrated for sites

where anemometer traces were available. Future research aimed at evaluating mathematical

models of hurricane wind speeds, wind directions, and decay rates is outlined.

96
INTRODUCTION

On the evening of September 12, 1979, Hurricane Frederic struck the Gulf Coast of

the United States between Pascagoula, Mississippi and Mobile, Alabama. Preliminary

reports of the damage indicated that while significant economic losses had been suffered,

structures hit by the hurricane were generally only partially demolished, expect along the

shoreline where the storm surge was heaviest. Consequently, it was felt that a visit to

the area for a survey of structural damage could provide information which would be useful

in evaluating current building codes and construction practice.

For the evaluation of building codes and construction practice to have significant

meaning, a knowledge of the wind speeds encountered is critically needed. In previous

studies of hurricanes which have caused significant damage, wind speed and wind direction

data have generally been quite limited and consequently estimates have been based largely

on "educated guesses" or estimated from the damage. Fortunately, a large number of anemommeters survived Hurricane Frederic and an extensive amount of data is available. Thus it

is possible to produce what is perhaps the most complete description available to date of

hurricane surface winds over land that is based on actual wind measurements.

The description of hurricane wind speeds and their spatial distribution which is

produced by this study has several potential applications in addition to the definition of

wind speeds for use in evaluating codes and building practice. A map of wind speeds during

Hurricane Frederic which is based largely on observations of damage has been prepared by

Fujita, Wakimoto and Stiegler [1] • The wind speeds determined from anemometer measurements

will provide a check of the estimates obtained from damage observations. Also mathematical

models have been developed for wind speeds associated with hurricanes and for the decay of

the storm over land. Two such models have been used in programs for predicting hurricane

wind speeds [2,3]. Future work is planned to evaluate the suitability of these mathematical models in describing the measured wind speeds and it is expected that it will be possible to make some estimation of the decay rate of Hurricane Frederic's winds as it moved

inland.

97
ANEMOMETER SITES AND CONDITION OF EQUIPMENT

Wind data from Hurricane Frederic was obtained from ten sites following the storm.

Four of the sites were near Mobile, Alabama and six were located along the coast between

Gulf port, Mississippi and Fort Walton Beach, Florida- The locations of the sites are shown

in Figure 1 together with their relation to the estimated storm track. Although none of

the stations were primary National Weather Service (NWS) stations with fastest-mile

recorders, four were airport locations and seven had strip chart wind recorders. A brief

description of the anemometer sites listing terrain type, instrumentation, height of

instrumentation and type of data available is given in Table 1. A classification of

terrain types is given in Table 2.

Three of the four airport stations could easily be classified as having open exposure,

type 2, conditions for the wind directions experienced during the passage of the hurricane.

However, the Pensacola Naval Air Station (PNA) site was not as clean. At PNA, the anemometer was mounted on the control tower roof and experienced an open exposure for winds

blowing from directions of 0 to 70 degrees and 185 to 360 degrees, measured clockwise from

North. For wind directions between 75 and 180 degrees winds passed over trees which ranged

from 20 to 40 feet (6-12 meters) in height and the terrain is assumed to be type 3 or

similar to that over the outskirts of towns.

Terrain exposures at several other sites could not be clearly classified as fitting a

specific type without any reservations. A brief discussion of the sites follows:

(a) Dauphin Island Bridge (DIB): The instrumentation is located on top of the

drawbridge super structure approximately 60 feet (18 meters) above the water level. It is

assumed that the exposure is type 1 (sea surface in strong winds), although Dauphin Island

forms a barrier to the south and southeast on the bridge. However, because of the height

of the instrumentation, the predominance of a sea exposure, particularly from the east, and

the fact that the drawbridge portion is at some distance from the main body of the island

or mainland, a sea exposure is considered appropriate.

(b) Gulf port Civil Defense (GCD): The instrumentation is located on a tower mounted

atop a 25 to 30 foot (8 to 9 meter) building near the center of Gulf port. The strongest

winds (only the peak wind speed was recorded) came from the north or northwest across a

98
large port ion of the city* The city is not heavily built up with large buildings, and,

consequently, the terrain is assumed to be similar to either the outskirts of towns or

centers of towns , type 3 or type 4

.

(c) Ingalls Shipbuilding (ISB): The exposure to the north and west is that of open

terrain with some small bodies of water. However, there are some large warehouse-type

buildings near the fire station where the anemometer was located. Winds from the south

and southwest would have a sea exposure but the site was obstructed by cranes and buildings. The assumption that the terrain exposure is type 2 (open terrain) was considered to

be reasonable, despite possible local effects due to neighboring buildings. During the

storm, a portion of the tower above the anemometer collapsed, the direction vane was

destroyed and the anemometer was tilted to the south at approximately 45 degrees. Fujita,

et al . [1] also noted this damage and a picture of the anemometer is shown in their report.

Wind tunnel tests on a similar anemometer tilted at 45 degrees indicated little change in

the velocity measurements and consequently it is suggested that the data not be changed due

to the tilt.

(d) Mobile Civil Defense (MCD) : The anemometer was located two-thirds of the way up

a water tower, (Figure 2) and was shielded for a number of wind directions by the tower

legs, struts and center pipe. The tower was located on a rise west of the center of

Mobile, about two miles into the suburbs. The city and the suburbs are heavily wooded with

the trees being 30 to 50 feet tall (9 to 15 meters). Terrain type 3 or 4 was assumed.

(e) Coast Guard Cutters: The "Salvia" was moored at the Alabama State Docks and was

shielded by a warehouse and a larger boat. Consequently, its wind data is highly questionable. The "White Pine" was located midstream in the south channel of the Mobile River

south of Twelve Mile Island. The river banks are well forested with pine trees having an

average height of 25 to 30 feet (7.5 to 9 meters). It was not possible to visit this site

but a detailed report was obtained courtesy of the Quarter Master on the White Pine. This

site was one of the most difficult to characterize. Information in Table 2 would lead to

an assumption of category 6, pine forest, but that assumption produced wind speeds which

were not consistent with any of the other data. Based partly on the results of analyzing

the data using different assumptions of terrain type, it was decided to use terrain types

2 or 3.

99
Most of the instrumentation at the various sites received regular service by either

military or NWS personnel and it can be assumed that the data are fairly accurate, indeed

as accurate as the data which forms the base for extreme wind speed estimates in other

parts of the United States. There are several exceptions- The anemometer at the MCD site

had not been serviced for a number of years but it was removed following the storm and

calibrated in the NBS wind tunnel by personnel from the Fluid Mechanics Division. Subsequently, data were corrected using the calibration curve. Also, the anemometers on the

Coast Guard cutters had reportedly not been serviced since installation.

STORM TRACK

An estimate of the storm track is shown in Figure 1. The track was determined on the

basis of radar pictures obtained from the Naval Weather Service at the Pensacola Naval Air

Station. One of the radar pictures is shown in Figure 3. The track is similar to the one

reported by Fujita, et al. [1]. It is recognized that the center of minimum pressure, the

center of the region of minimum winds and the center of the rainless portion in the radar

pictures may not coincide. However, in the absence of the detailed pressure and velocity

information required, the center of the rainless portion of the radar picture provides a

useful approximation of the location of the eye of the storm.

ANALYSIS OF WIND DATA

In order to compare the wind speed data obtained from Hurricane Frederic with design

wind speeds, it is necessary to express them in the form of fastest-mile wind speeds in a

standard exposure. Normally, such design wind speeds are given for open exposure conditions and a height of 30 feet (10 meters) [2,4]. Consequently, it is necessary to transform the data from the exposure and height corresponding to the actual environment of the

instrument to standard exposure and height conditions, and from a mean or gust value to a

fastest-mile value.

Procedures for transforming from one averaging time to another, from one terrain to

another and from one elevation to another are contained in Ref. [5]« The transformation

for different averaging times is based on work by Durst as modified by Simiu and Scanlan

[5]. The equation is as follows:

100
U

t (z) u 3600 (z) i + 0.98 c(t)

£n (z/z Q

) (1)

where U t (z) velocity for averaging time t in seconds at height z

u 3600 (z) mean hourly velocity at height z

o roughness parameter for side exposure

c(t) coefficient obtained from table 3, linear interpolation is allowed.

This procedure had not been verified for winds in hurricanes. Consequently, initial work

was performed to check the procedure. The check was accomplished by determining peak to

mean velocity ratios for 20 to 30 minute segments of the wind records from the seven

anemometer sites with strip chart recordings. The peak to mean velocity ratios represent

the effect of averaging time, between 2 second averages and 1200 second averages, on the

peak velocity estimates. Probability distributions of the peak to mean ratios are plotted

in Figure 4 for the seven sites which had strip chart records. Corresponding ratios based

on Durst 's work as modified by Simiu and Scanlan are plotted at the 50 percent probability

level. In general, the agreement is quite good except for the data from MCD where interference of the water tower may affect the data, see Figure 2. Based on this comparison, it

was determined that Eq. (1) and the values of c(t) in Table 3 could be used with hurricane

wind data.

Procedures for transforming wind speeds from one elevation to another and from one

exposure to another are also contained in Ref . [5]. The transformations are based either

on the power-law or the log-law and are designed to be used with mean hourly wind speeds.

The power-law transformation requires knowledge of the power-law exponent, a, and the

boundary layer height, 6, for each terrain in question. The log-law transformation

requires knowledge of the roughness length parameter for each terrain. Table 2 contains

values of these parameters for the various types of exposure assumed. The values are

obtained largely from Tables 2.2.1 and 2.2.2 of Ref. [5] but also rely on input from

Ref. [6] and [7J. Judgment was used in order to produce an even gradation of exposure

The log-law and power-law relationships may not apply to hurricanes, especially near

the eye wall where winds are highest. Furthermore, even if they do apply there is no

types

.

guarantee that the parameters developed from data for other types of storms are applicable

101
to hurricane winds- However, the differences between the anemometer elevations and the

standard reference elevation are generally small and the approximations involved are likely

to be acceotable

.

Available mean, one minute average, and gust velocity data were transformed to

equivalent mean, gust, and fastest-mile wind velocities for open terrain conditions at 10

meters. Initially, all data was transformed to equivalent mean hourly values. Next, the

mean hourly values were transformed from the specific terrain and elevation of the instrumentation to the standard exposure conditions. Finally, the hourly values were transformed

to gust and fastest-mile values of the standard exposure conditions. Consistency of the

transformations was checked by recalculating the peak to mean velocity ratios for the

transformed data. To obtain the peak to mean ratios, gust velocities for standard exposure

conditions which were obtained from original gust velocity data were divided by corresponding mean velocity data for standard exposure conditions which were obtained from the

original mean velocity readings.

Peak to mean velocity ratios for corrected data at the airport locations with the best

defined exposure conditions, i.e., MOB, PRA, and EAF, are shown in Figure 5. The equivalent coefficient from Durst's relationship is also shown. Roughly eighty percent of the

peak to mean ratios observed were lower than Durst's value. Figure 6 shows similar probability distributions for peak to mean velocity ratios at the other sites and clearly shows

that the transformation is causing the data to collapse toward expected values for open

terrain, (solid curve in Figure 5) • The only curve which does not collapse well is the

one corresponding to MCD data. As noted earlier, the MCD site may suffer from interference

effects

.

Transformations of the one minute average readings involved an additional step because

the data does not represent the maximum one minute average during a certain period of time.

Rather, the one minute average reading is taken at an arbitrary time, usually every hour.

Consequently, a transformation which uses Eq. (1) and a c(t) corresponding to 60 seconds

would be misleading since Eq. (1) refers to maximum values. A relationship between the

arbitrary one minute average reading and the fastest-mile reading during that hour has

been suggested by Thom [8] and is expressed as follows:

U fm = 9.55 + 0.999 Um (2)

10Z
where Ufm = fastest-mile wind speed during an hour

Um

= one minute average reading during the same hour.

Consequently, one minute average readings were first transformed to fastest-mile wind

speeds which were then transformed to mean hourly wind speeds using Eq. (1).

RESULTS OF DATA ANALYSIS

The results of the data analysis are presented in Figures 7 through 14 as fastest-mile

wind speeds during hour intervals. Results are presented for 8 of the 10 sites for times

between 1700 hours Greenwich mean time (GMT) on September 12 and 1500 hours GMT on

September 13. Data from MCD is not presented because of the questions concerning its

quality. The data from GCD consisted of a single peak wind speed observation with the

corresponding wind direction.

Note that the fastest-mile wind speeds predicted from the different types of data,

including one minute average readings, are quite consistent. This suggests the validity

of the data analysis procedures. Different types of terrain conditions were assumed for

each of the sites in order to determine the effects of overestimating or underestimating

the terrain roughness. Changing the exposure type to the next rougher type could increase

the fastest-mile speed from 12 to 25 percent depending on the site. However, the assumption of a terrain category other than the one chosen also resulted in a discrepancy between

the fastest-mile wind speeds predicted from the gust speed on the one hand and the mean

velocity data on the other hand

.

FUTURE WORK

Analysis of the wind speed and direction data from Hurricane Frederic is continuing.

Future work includes evaluation of the consistency of the wind speed and direction data by

comparing measured values with corresponding radar pictures such as the one shown in

Figure 3. A major emphasis will be placed on attempting to fit the mathematical models

used in Refs. [2] and [3] to the wind and direction data. These models may provide useful

tools for obtaining wind speeds between the anemometer locations and provide a basis for

determining the rate of decay of the storm as it moved inland.

103
REFERENCES

Fujita, T., Wakimoto, R. , and Stiegler, D., "Mesoscale Damage Patterns of Hurricane

Frederic in Relation to Enhanced SMS Imagery," Proceedings of the 19th Conference on Radar Meteorology, Miami Beach, Florida, April 15-18, 1980.

Batts, M., Cordes, M., Russell, L., Shaver, J., and Simiu, E., Hurricane Wind Speeds

in the United States , Building Science Series Report 124, National Bureau of

Standards, Washington, D.C., May 1980.

Tryggvason, B., Surry, D., and Davenport, A., "Predicting Wind-Induced Response in

Hurricane Zones," Journal of the Structural Division , ASCE, Vol. 102, No. ST12,

December 1976, pp. 2333-2350.

"Building Code Requirements for Minimum Design Loads in Buildings and Other

Structures," (ANSI A58. 1-1972), American National Standards Institute, New York, 1972

Simiu, E., and Scanlan, R. , Wind Effects on Structures , Wiley-Interscience , New York,

N.Y., 1978.

Counihan, J., "Adiabatic Atmospheric Boundary Layers: A Review and Analysis of Data from the Period 1880-1972," Atmospheric Environment , Vol. 9, 1975, pp. 871-905.

Engineering Sciences Data Unit, "Characterstics of Wind Speed in the Lower Layers of

the Atmosphere Near the Ground: Strong Winds (Neutral Atmosphere)," ESD Item No. 72026, London, England, November, 1972 (with Amendments A and B, October, 1974).

Thorn, H.C.S., "Prediction of Design and Operating Velocities for Large Steerable

Radio Antennas," Large Steerable Radio Antennas - Climatological and Aerodynamic

Considerations, Annuals of the New York Academy of Sciences, Vol. 116, Art. 1, pp. 90-100, New York, N.Y., 1964.

104
Table 1. Characteristics of Anemometer Sites

Types of Data Available

Site TerrainType(l) Height of Anemometer

Type of InstrumentationStrip

i I, i i i Peak Observation

1 Avg Minute

Readings

Dauphin Island Bridge (DIB) 1 ^18m 3-cup and vane, NWS type Yes No No

Eglin Air Force Base (EAF) (airport)2 4 . 6m 3 blade Impeller and aerovane

Yes Yes Yes

Gulfport Civil Defense (GCD) 3 or 4 -v27m 3-cup and vane, NWS type No Yes No

Ingalls Shipbuilding (ISB) 2 "v-lOm 3-cup and vane, NWS type Yes No Ho

Mobile Civil Defense (MCD) 3 or 4 •v23m 3-cup and vane, NWS type Yes No :,

Mobile NWS, Bates Field (MOB) (airport)2 6.7m 3-cup and vane, NWS type Yes Yes Yes

Pensacola Naval Air Station

(PNA) (airport)

2 and 3(2) x23m impeller and aerovane Yes Yes V e a

Pensacola Regional Airport

(PRA)

2 6. 7m 3-cup and vane, NWS type Yes Yes Yes

Coast Guard Cutter White Pine

(near Twelve Mile Island)

2 or 3 %15m above water level Impeller and aerovane No Yes YeS

Coast Guard Cutter Salvia

(at Alabama State Docks) shielded M5m above water level impeller and aerovane No No Yes

(1) See table 2 for description of terrain type.

(2) Type 2 for wind directions 0-70 and 185-360, type 3 for wind directions 75-180.

Table 2. Velocity Profile Characteristics

Exposure

Type Type of Surface

Log-Law Surface Roughness Parameters

z (m)

Power-Law

Exponent

Nominal Boundary

Layer Height for Use with Power-Law

6 (m)

0 Sand .0005 .10 225

1 Sea surface (high winds) .005 .12 250

2 Grass (open) .05 .15 275

3 Outskirts of towns .20 .19 340

4 Centers of towns .40 .22 400"

5 Centers of large cities . 70 .26 430

6 Pine forest 1.00 .30 460

7 Extremely high roughness 2.00 .35 500

Table 3. Coefficient c(c)

t 1 10 20 30 50 100 200 300 600 1000 3600

c(t) 3.00 2.32 2.00 1.73 1.35 1.02 0.70 0.54 0.36 0.16 0.00

105
106
Figure 2. Anemometer Site at Mobile Civil Defense (MCD).

107
Figure 3. Radar Picture of Hurricane Frederic at 0315 GMT.

108
o © o © o

o

~1 1

A 0 g

A <E O

O

O D O

T

O

oo

SO

o

0 j# 0

4 o

o ^ •

A o o a aO °

-'o

o

©

o

o ©

©

©

oe

a

a

Aa <*iQ

9

O A

if

9

O

v

V

Typo of torrain V atsumtd & height Ratio aftor Location of anemometer Durst O EAF 2 4.1m 1.63, •

PfiA 2 8.7m 1.59, V A ISB 2 -10m 1.54, * OMOB 2 6.7m 1.58, V O DIB 1 - 18m 1.35, •

V VMCO 4 -23m 1.71, • PNA1 2 23m 1.47, * © PMA2 3 23m Uti I

1

- For wind titroctiMi 0 70 mi 1M-34Q

2 - For wind *iroctkM« 75-180

1.0 1.1 1.2 1.3 1.4 1.5 1.6 1.7 1.8 PEAK TO MEAN VELOCITY RATIO

1.9 2.0 2.1 2.2

Figure k. Probability Distributions of Peak to Mean Velocity Ratios

from Actual Data.

109
16-0

90 -

80

70

(0

| SO

EJCDQ

£ 40

30

n

i

1 1 r

—»01!ST COEFFICIEDT FOR OPEN TERRAIN

WIND SPEED DATA CONVERTED TO )• IN •PEN TERRAIN Z e

= 1 05m O MOB TERRAIN TYPE 2 A PtA TERRAIN TYPE 2 LAP TERRAIN TYPE 2 COMPOSITE

1 2 t 3 1.4 1.5 IS 1.7 II

PEAK TO MEAN VaOCITY RATIO

1 9 20

Figure 5» Probability Distributions of Peak to Mean Velocity Ratios

at Three Airport Locations - Transformed Velocities.

too

E

o

O

WIND SPEED DATA CONVERTED

TO t*n IN OPEN TERRAIN. U = • 05m O PNA TERRAIN TYPE 2 WIND DIRECTIONS 0-70 AND 185 160 A PNA • TERRAIN TYPE 3 WIND DIRECTIONS 75-180

ISB - TERRAIN TYPE 2 O WB -TERRAIN TYPE 1 O MCD TERRAIN TYPE 4 —COMPOSITE FOR MOB,

P«A. EAF

1.2 1 3 1.4 1.5 It 1.7 18 PEAK TO MEAN VELOCITY RATIO

1 9 20 21

Figure 6. Probability Distributions of Peak to Mean Velocity Ratios

- Transformed Velocities.

110
140

SO

3 60

1 1 1— 1 III! TERRAIN TYPE 2i

ckctect uii t u/iun CBrcnc rAiltol MILt WINU arttUa ESTIMATED FROM: A 20 to 30 MIN AV6

DIRECTION • PEAK 60STS 1 -

A

o f

: o

- A• * A

A 0 •

• • A

A 0 ± *

o •

-• . • •

111 OF STORM • •

1 1 1 1 1

l l l 1 l l 1700 1900 2100 2300 100 300 500

TIME [GMT)

700 BOO 1100 1300 1500

320

340

0

20

40

60

80

100

120

140

160

180

200

220

240

Figure 7. Maximum Fastest Mile Wind Speeds During Hour Intervals at ISB.

140

120

S5 80 o

z

*

3 60

5 40

20

DIRECTION

FAST MILE WIND SPEEDS

ESTIMATED FROM: A 20 to 30 MIN AV6. • PEAK GUSTS

1700 1900 2100 2300 100 300 500

TIME |6MT)

700 900 1100 1300

320

340

0

20

40 1—

10

10 l

100 O

X. 120 o

h-O

140

u_> Cca

160 a

s

ISO *

200

220

240

1500

Figure Maximum Fastest Mile Wind Speeds During Hour Intervals at DIB,

111
0 I I I I I _l 1 I I 1 - I U 240

1700 1900 2100 2300 100 300 500 700 900 1100 1300 1500 TIME (GMT)

Figure 9- Maximum Fastest Mile Wind Speeds During Hour Intervals at MOB.

Figure 10. Maximum Fastest Mile Wind Speeds During Hour Intervals at PNA.

112
0 I 1 1 1 1 I I I I I I U 240

1700 1900 2100 2300 100 300 500 700 900 1100 1300 1500 TIME [6MT|

Figure 11. Maximum Fastest Mile Wind Speeds During Hour Intervals at PRA.

Figure 12. Hourly Fastest Mile Wind Speeds from CGC White Pine at 12 Mile Island.
gure 13. Hourly Fastest Mile Wind Speeds from CGC Salvia at Alaoama State Docks

Figure lh. Maximum Fastest Mile Wind Speeds During Hour Intervals at EAF.

114
HURRICANES DAVID AND FREDERIC 1979

by

Celso S. Barrientos

The 1979 tropical cyclone season indicated some signs of returning to a "normal"

situation. These normal characteristics are lower latitude origins and tracks of tropical

cyclones, more United States hurricane landfalls and more intense hurricanes hitting land

areas in the Western North Atlantic. This "normal" behavior of hurricanes produced the

greatest annual combined damage in the United States from tropical cyclones.

Considering equity among the sexes and human rights, 1979 was the first year male

names were used for the western North Atlantic hurricanes. Female and male names were

used alternatively in an alphabetical list. Although the chance for this happening is

relatively small, the two most destructive hurricanes in 1979 were male —hurricanes David

and Frederic.

Hurricanes David and Frederic were the strongest pair of storms to occur in the North

Atlantic in a decade. In 1969, hurricanes Camile and Debie were the last two such powerful

storms that roamed in these waters.

David and Frederic were like brothers —"normal" type hurricanes. Both may be regarded

as Cape Verde type. The characteristics are: (1) a seed storm originating from the Cape

Verde area; (2) a track of parabolic character around the periphery of the Azores-Bermuda

high; (3) an intensity maintained for a long duration unless weakened by landfall; and

(4) a size which expands in moving to higher latitudes, and before reaching the United

States. Both hurricanes survived for more than two weeks and both caused billion-dollar

damage

.

David was probably the most intense hurricane of this century in the eastern

Caribbean Sea area. It reached a maximum strength of 150 kt and minimum central pressure

of 924 mb south of Puerto Rico on August 30, and its strength did not diminish when David

struck Dominica and Santo Domingo. It was the strongest hurricane reported at Dominica

since 1834 and at Santo Domingo since 1930. Fortunately, David was no longer a major

hurricane when it struck the United States. The landfall was just north of Palm Beach,

Florida, around mid-day September 3. The central pressure during landfall was 972 mb

.

Estimated maximum winds were 85 kt. There was little intensity change before the second

115
landfall near Savannah Beach, Georgia, approximately 24 hours later. Savannah station

reported a minimum pressure of 970 mb. Since David was a big hurricane, it affected the

whole east coast of the United States.

David left a death toll of 56 in the island of Dominica with 60,000 of the 80,000

residents homeless. There were seven deaths in Puerto Rico and damage estimated as 70

million dollars. The government of the Dominican Republic estimated the death toll there

in excess of 1,200 with damage exceeding 1 billion U.S. dollars and 200,000 left hornless.

In the United States, five deaths were directly attributed to David and total damage

exceeded 300 million dollars.

Frederic formed in a manner similar to David. It appeared that a second Cape Verde

hurricane would shortly move through eastern Caribbean area while David was still wreaking

havoc further west. Fortunately, the strength of David caused the weakening of Frederic

as the warm outflowing air aloft from David descended onto Frederic and prevented development. Frederic regained strength when it reached the Gulf of Mexico. Hurricane warnings

went into effect from Grand Isle, Louisiana, to Panama City, Florida, at 0930 p.m. CDT

September 11.

Frederic became the first hurricane to strike Mobile, Alabama, directly since 1926.

The central pressure of 946 mb and estimated maximum winds of 115 kt at landfall made

Frederic the most intense hurricane of this century to affect the Mobile-Pascaguoula

,

Mississippi, area. The highest wind reported in the U.S. was a gust of 126 kt on the

Dauphine Island, Alabama, bridge while a gust of 119 kt was observed at the Dauphine

Island Sea Lab before the equipment was destroyed. The peak storm surge of 12 feet over

Gulf Shores, Alabama, destroyed much of the island. Five deaths have been directly

attributed to Frederic. The estimated damage is $2.3 billion. This exceeds the $2.1

billion in damage attributed to Agnes in 1972, and makes Frederic the costliest United

States hurricane in history though no adjustment was made for inflation.

116
SEISMIC RESPONSE OF FLOATING OFFSHORE STRUCTURES

by

Hajime Tsuchlda

Susumu Iai

Setsuo Noda

ABSTRACT

A study was made on the technical feasibility of a floating offshore structure from

viewpoint of seismic performance. The structure selected for the study mainly consists o

a pontoon and 37 dolphins. The pontoon is 5 km in length and 0.84 km in width and it is

connected to the dolphins by links and dampers. The damper has a non-linear force deformation relationship to limit the force to the dolphin below an acceptable level. For the

seismic response calculation, a time history of earthquake ground motions covering a very

wide period range was needed and a procedure to synthesize such a time history was developed. The seismic response to the synthesized ground motions was calculated. It was

found that the structure seems to be feasible for seismic performance.

117
INTRODUCTION

Recently many floating offshore structures have been proposed as a fishery base,

storage faciity, airport, and so on. [1,2] In the case of floating offshore structures,

a pontoon which may also be called a floating body is moored at a position by a chain and

sinker mooring system, a tension leg mooring system, or a dolphin and link mooring system.

A study was made to examine the technical feasibility of a large floating offshore structure using the dolphin and link mooring system from a viewpoint of seismic performance.

In the study a method to synthesize a time history of earthquake ground motions of wide

period range was developed, and an earthquake response of a sample offshore structure to

the synthesized motions was calculated. This paper presents the method of synthesis and

results of the response calculation.

The floating offshore structure selected as an example in the study mainly consists

of a pontoon 5 km in length and 0.84 km in width in plan and 37 dolphins to which the pontoon is connected by the links. The link is connected to the pontoon through a tubular

rubber fender and this system is called a damper in this paper. The damper is provided to

limit a force in the link below some acceptable level.

The large mass of the pontoon and the spring effect of the damper and the dolphin form

an oscillation system with a very long natural period of more than 15 seconds. As input

earthquake ground motions to this structure, none of the strong-motion earthquake records

available at present are appropriate because the strong-motion accelerograph can not record

components of such long period with adequate accuracy. The dolphin itself is sensitive to

rather short period components of earthquake ground motions. This is the reason a method

to synthesize earthquake ground motions needed to be developed.

It was difficult to assess the earthquake response of the proposed floating offshore

structure before the study was made, so the response calculation was computed using synthesized, earthquake ground motions. For the response calculation a number of considerations

were involved in organizing the calculation system.

FLOATING OFFSHORE STRUCTURE

The plan view of the selected structure is shown in Figure 1. It consists of a

pontoon of 5 km in length and 0.84 km in width and 37 dolphins. The details of the

118
dolphin and the damper are given in Figure 2. The damper has a non-linear force

deformation relationship as can be seen in Figure 3. The damper functions in the following

way. When the external forces to the pontoon are not too large, the pontoon is fixed somewhat rigidly and its oscillation is kept small. When rather large relative displacement

takes place between the pontoon and the dolphin during an earthquake, the force in the

link is kept below an acceptable level so that the force to dolphin is also kept below the

level. The soil conditions at the structure site are considered roughly those shown in

Figure 2.

When the dolphin is considered to be rigid, the total structure has a natural period

of about 18 seconds for sway oscillation. The dolphin, when it is separated from the pontoon, has a natural period of about 1 second. Therefore, it was considered that, for some

parts of the structure such as the link and the dolphin, both the long period components

and the short period components of earthquake ground motions might be significant. Since

the structure employs dampers which show distinct non-linearity, the mode superposition

technique is not applicable. So, a time-history of earthquake ground motions covering a

very wide period range was synthesized, and the response of the structure was calculated

against the synthesized ground motions.

INPUT EARTHQUAKE MOTIONS

The Synthesis of Input Earthquake Motions of a Wide Period Range

Input earthquake motions are usually obtained in two ways; one is to use observed

earthquake motions of which amplitude is adjusted for the condition under consideration,

and the other is to simulate earthquake source mechanism and wave propagation.

Earthquake motions have been observed by different kinds of seismographs. The period

band covered by a seismograph corresponds to the frequency characteristics of the instrument. For example, a SMAC-B2 accelerograph covers a period band from 0.1 second to several

seconds, the one-magnification strong-motion seismograph deployed by the Japan Meteorological Agency covers a period band from one second to ten seconds, and the Lp-Lo type seismograph deployed by the Agency covers a period band from 10 seconds to 50 seconds. There is

119
almost no seismograph which covers a period range longer than 50 seconds- Moreover, there

is no seismograph which covers a wide period range, for example a range from 0.1 second to

00 seconds.

Theoretical seismograms are obtained using several different kinds of earthquake

source mechanism and wave propagation models. Period range covered by the model depends on

the complexity of the model. For example, fault models which are considered simplest [3]

cover a period range from several ten seconds to 00 seconds. Although the number of parameters required for this model is not small, most of the parameters can be obtained. There

are several models which are more complex and cover a wider period range [4]. Information

currently available on the earthquake source mechanism is, however, not enough to determine

the parameters of these complex models.

In order to cover a wide period range from 0.1 second or short to 00 seconds, a

synthesis method of input earthquake motions of a wide period range was proposed by the

authors [5]. The method is as follows: The wide period range is divided into four bands;

one from 0.1 to 1-2 seconds, one from 1-2 to 6-10 seconds, one from 6-10 to 30-50 seconds,

and one longer than 30-50 seconds. The borders between the bands are not sharp and they

are expressed as period ranges

.

The input earthquake motions are synthesized from four components of earthquake ground

motions corresponding to these four period bands. The component of the period band longer

than 30-50 seconds is theoretically calculated by a fault model and the other three are

formed by adjusting the amplitudes of seismograms of actual earthquake motions observed

by three types of seismometers. The amplitudes are adjusted according to empirical formulae on relationship among maximum recorded amplitude, magnitude, and distance to epicenter

or fault plane from observation point. The empirical formulae are: the formula of surface

wave magnitude [6], Tsuboi's formula [7], and Noda and Uwabe's formula [8], respectively.

Each component is filtered by a corresponding band-pass filter after adjusting the amplitudes. Initial time of each component is adjusted so as the arrival times of the S waves

of each component coincide. Finally, these four components are summed up to form input

earthquake motions. It is believed that the earthquake ground motions synthesized in this

way are reliable and achievable considering the present state-of-the-art in earthquake

engineering and seismology.

120
Phase Delay of Input Earthquake Motion on a Base Layer

As the dolphins are located over a wide area, the time lag of seismic wave arrival

among the locations of the dolphins has to be considered. In the study, seismic waves are

considered to propagate in 3.5 km/s in a base layer straight from the epicenter to the

structure location. The velocity 3.5 km/s is considered, in seismology, to be the velocity of S-wave propagation through the earth crust. Although actual seismic wave propagation in a base layer may be complex, this treatment of phase delay is a practical and

acceptable conception in engineering practice.

Wave propagation has been observed by two of the authors using a two dimensional

seismometer array which extends 2500 meters [9,10]. It is hoped that analysis of the

data from this observation will provide further information.

Synthesis of Input Earthquake Motions

One earthquake considered in this study is of magnitude 8.1. Parameters of this

earthquake and other data used for the synthesis are summarized on Table 1. Figure 4

shows the ground motions of four period bands and the synthesized motions. Figure 5

shows the input earthquake motions synthesized for the response analysis.

RESPONSE CALCULATION

The response calculation was made using the procedure shown in Figure 6. The surface

layer overlying the design base level is idealized into a lumped mass system with nonlinear springs which represent the soil stiffness, and the response of the system to the

synthesized ground motions was calculated. This response was considered as input motion

to the dolphins. The dolphin without the linkage to the pontoon was idealized into a

lumped mass system as shown in Figure 7 and response was calculated. Stresses in the

piles due to the earthquake response were next calculated. The modal analysis of the

idealized system was made and the natural periods of the modes were estimated.

In the earthquake response calculation of the total structure, each dolphin was

idealized into a single mass system. If each dolphin had been ideaxized as the model in

Figure 7, the model of the total structure with 37 dolphins would have too many elements

121
to be accomodated in a computer of a practical, available size. For this reason the

dolphin was idealized into the single mass system. The idealization was based on the

result of a model analysis of the dolphin.

The mass of the idealized model representing the mass of the dolphin was connected

to the pontoon with a spring which represented the force deformation relationship of the

damper in Figure 3. The pontoon was assumed to have freedom of displacement in the x-axis

direction (sway), in the y-axis direction (surge), and rotation about the z-axis (yaw).

The model of the total structure for the response calculation is shown in Figure 8.

Damping by sea water was estimated from the numerical simulation [11]

when the dolphin was idealized into the single mass system, interaction effects

between the pile and the surface layer could not be taken into the response calculation.

Therefore, for calculating stresses in the piles, the stresses calculated from the earthquake response of the dolphin without linkage to the pontoon and the stresses due to the

force in the link calculated in the earthquake response of the total structure were superposed. In principle, this is not accurate; however, it may be acceptable from a practical

point of view.

Accelerations, velocities, and displacements of the pontoon are shown in Figures 9,

10, and 11, respectively. Deformation and force of one of the dampers (at the connection

between the No. 11 dolphin and the pontoon) are shown in Figure 12.

From this study, the floating offshore structure appears feasible from a seismic

performance point of view. Further details of this study are available in separate

papers [5,12]

.

CONCLUSIONS

(1) A procedure to synthesize the time history of earthquake ground motions covering a

very wide period range was developed and presented.

(2) A procedure to calculate the earthquake response of a floating offshore structure

which consists of a pontoon, dolphins, links, and dampers with non-linear force deformation

relationships was organized and presented.

(3) From the response calculation, the floating offshore structure considered in this

study was found to be technically feasible from a seismic performance viewpoint.

122
REFERENCES

[1] Kato, W., "Planning of Offshore Building-Project '76 Offshore Fishery Base," Column,

No. 66, October 1977, pp. 29-36.

[2] Offshore Airport Study Group, "Floating Offshore Airport in Osaka Bay, Japan, Digest

of Preliminary Engineering Study," Proceedings of the First International Conference

on Offshore Airport Technology, Bethesda, Maryland, 29 April-2 May 1973, Vol. 2, pp. 43-49.

[3] Haskell, N. A., "Elastic Displacement in the Near-field of a Propagating Fault,"

Bulletin of the Seismological Society of America, Vol. 59, No. 2, April 1969,

pp. 865-906.

[4] Wiggins, R. A., Sweet, J., and Frazier, G. A., "Seismic Risk for Offshore

Structures," 10th Annual Offshore Technology Conference in Houston, 1978,

pp. 529-533.

[5] Iai, S., and Tsuchida, H. , "A Synthesis Method of Input Ground Motions of a Wide

Period Range," Report of the Port and Harbour Research Institute, Vol. 19, No. 1,

March 1980, pp. 63-96.

[6] Utsu, T., "Seismology," Kyoritsu Publisher, January 1978.

[7] Tsuboi, C, "Determination of the Gultenberg-Richter ' s Magnitude of Earthquakes

Occurring In and Near Japan," Jishin, II, Vol. 7, No. 3, 1954, pp. 185-193.

[8] Noda, S., Uwave, T., and Chiba, T., "Relation Between Seismic Coefficient and

Ground Acceleration for Gravity Quaywall," Report of the Port and Harbour Research

Institute, Vol. 14, No. 4, December 1975, pp. 67-111.

[9] Hayashi, S. and Tsuchida, H. , "Observation and Analysis of Ground Response in

Earthquakes," Proceedings of the Sixth Joint Meeting of U.S. - Japan Panel on Wind

Seismic Effects, U.S. -Japan Natural Resources Conference (UJNR), May 1974.

[10] Tsuchida, H., Kurata, E., and Hayashi, S., "Observation of Earthquake Response of

Ground with Horizontal and Vertical Seismometer Arrays," Proceedings of the Sixth

World Conference on Earthquake Engineering, Vol. 1, January 1977, pp. 509-514.

[11] Ueda, S. and Shiraishi, S., "On the Characteristics of Oscillations of Multi-Cylinder

Floating Structure by Numerical Simulation, " Technical Note of the Port and Harbour

Research Institute, No. 339, July 1980 (in press).

[12] Iai, S. and Tsuchida, H., "Earthquake Response Analysis of Floating Type Structure,"

Technical Note of the Port and Harbour Research Institute, No. 337, July 1980 (in

press)

.

123
Table. 1 Procedure of synthesis of the earthquake motion of a wide period range

Period Band Longer than

30-50 s From 6-10 s to 30-50 s From 1-2 s to 6-10 s

' —

From 0.1 s to 1-2 s

Formation of

seismic wave

Simulation by

a faul t model

Actual seismic wave observed

by Lp-Lo type

seismometer of

J.M.A.

Actual seismic wave observed

by LMl type

seismometer of

J.M.A.

Actual seismic wave observed

by SMAC-B2 type

strong-motion

accelerograph

Determination

of peak value

Simulated wave

itself

Formula of

surface wave

magnitude^)

Tsuboi 's formula?)

Noda and Uwabe's

formula^)

Earthquake

of the

record wave

used for the

formation of

the seismic wave

1977.12.24.06.02

Epicenter 39°04'N

143°38'E

0 km

Magnitude 5.9

Observation station:

Matsushiro

1978 Miyagi-ken

Oki Earthquake

1978.06.12.17.14

Epicenter

38.09°N

142.12°E

40 km

Magnitude 7.4

Ofunato

rau 1

length 120 km

width 80 km

dip 10°

strike N40°E

Surface wave Magnitude

Ms= 8.2

Magnitude

(J.M.A.)

Mj= 8.1

Magnitude

(J.M.A.)

Mj= 8.1

Dislocation

Parameters

of the

assumed

earthquake

3.1 m

N50°w

Rise Time 9.2 s Rapture Velocity

3.0 km/s

Rapture

direction

N50°W

Vp 6.0 km/s

Vs 3.5 km/s

Epicentral

Distance

155 km

Epicentral

distance

155 km

Distance to the fault

plane

81 km

5000 m

o

co Pontoon

Plan view

1 Plan vi ew of floating offshore structure under study

1

Q

3

Dol phin

124
Fig. 2 Dolphin and damper

125
Deformation (cm)

Fig. 3 Load-deflection relationship of damper

126
cm

30. !-

„ cm

S.

20. cm

20.

0. 7 cm

Component of 30-50 seconds or longer max=29.0 cm

Component from 6-10 seconds to 30-50 seconds max=7.7 cm

Component from 1-2 seconds to 6-10 seconds max=17.6 cm

Component from 0.1 second to 1-2 seconds max=0.7 cm

-0. 7 cm

50. r Synthesized displacement max=46.9 cm

300. s

Fig. 4 Ground motions for four period bands and synthesized motions

127
60.

D.

cm

-60. L

N40W Displacement

30

cm

0

E40N Displacement

-30. L

Gal

300.

N40W Acceleration

0. ^-'•SmTTTI^

''

-'

''

1 '"' v'"

'l/

'sf,AlV—

IW" '

300.

300. r

-300

E40N Acceleration

50. 00.

max=50.9 cm

max=25.0 cm

max=300 Gal

max=236 Gal

150. 200. 250. 300. s

Fig. 5 Input earthquake motions for reponse calculation

128
Ideal i zation

of ground

>

Ideal i zation

of dolphin

(without link)

Response

calculation

(Input motions

to dolphin)

i

Synthesis of

ground motions at

design base level

T

Earth

respon

dolp

(withou

quake

se of

ni n t link)

Stre

in p (without

of pon

sses

i les

effects

toon)

Modal

analysis

L. „

Ideal i zation

of dolphin into

single mass

system

Idealization of

total structure

Estimation of

load to dolphin

Earthquake

response

calcul ation

Stresses

in piles

(with effects

of pontoon)

Stability

examination of

pontoon

Fig. 6 Flow of response calculation

129
15 dolphins (No. 23 - No. 37)

Fixed in absolute

coordinate

Damping

element

for yaw

No. 23

No. 22

Damper

(non-linear hysteretic

force-deformation

relationship)

Top floor of dolphin

1

No.l L^Damping

f-p element

I for surge

\ Steel pipe pile

130

Fig. 8 Idealization of floating offshore

structure under study

'22 dolphins

(No.l - No. 22)
Absolute acceleration of the floating body (N40W)

4L

u

0. 000Q09 r

0

•0. 000QQ9

rad/s* Absolute acceleration of the floating body (rotation)

1 I V

1

1 I 0. 50. 100. 150. 200.S

Fig. 9 Acceleration response of pontoon

Absolute velocity of the floating body (N40W)

cm/s

15 rcm/s Absolute velocity of the floating body (E40N)

0.

-15

rad/s

Absolute velocity of the floating body (rotation)

50. 100. 150.

Fig. 10 Velocity response of pontoon

200. s

131
13?
CRITICAL EXCITATIONS OF STRUCTURES

by

R. F. Drenick

F. Novomestky

G. Bagchi

INTRODUCTION

This paper is, in a way, a progress report on the development of a new method for the

assessment of structural reliability under dynamic (and notably under earthquake) loads.

The main advantage of the method is that it leads to results which are valid on a rather

high level of confidence. Such levels are often achieved at a price of excessive conservatism. The method avoids this by extrapolating from pertinent experience (concerning

e.g., strong ground motion) to all other excitations that can realistically be expected

at the site of interest. Among these excitations, it then determines those which are least

favorable to the structure. The determination has some characteristics which are unusual

in structural dynamics. Among them is the fact that it requires nonlinear programming

methods for its solution. The procedure has been called the "critical excitation method."

The high level of assurance to which the critical excitation method aspires can be

achieved by avoiding the use of information, particularly of a probabilistic kind, which

is not available on a comparable level. The paper first develops the line of reasoning

which indicates the need for this avoidance, then shows that the need leads to the method

in a rather natural way. It proceeds to give a brief account of the problem that must be

solved, and of the way in which it is being solved, as of this writing. Some typical

results are quoted from a recent assessment of the earthquake resistance of a nuclear

power plant in the Southeastern United States. The paper concludes with a discussion of

some of the advantages and disadvantages of the method.

133
Most of the discussion in the paper tacitly assumes that the problem of interest is

an assessment of structural reliability under dynamic loads due to strong ground motion.

Similar reliability problems however are equally good candidates for the treatment by the

method.

NON-ROBUSTNESS PROBLEMS

It has been pointed out in the preceding section that an assessment of the earthquake

resistance of a structure should avoid information which is not available on a high confidence level, high enough at any rate to be comparable with the desired reliability of the

structure. Such information is rather scant on strong ground motion. Vanmarcke [2]

suggests that the following items might qualify.

(a) Duration of strong ground motion

(b) Ground motion intensity, its peak value as well as its variation with time

(c) Envelope of the amplitude spectrum

(d) Effects of macro-zone, micro-zone and site soil characteristics

(e) Effect of focal distance

To these five items, one can add a sixth, namely

(f) Bounds on the peak ground acceleration and velocity.

Some of this information is quite specific but some of it is rather qualitative (and

none of it, one might note, is of a clearly probabilistic nature). One can perhaps use it

to identify excitations which are patently inconsistent with all current seismic theory and

experience, and which therefore are not realistic candidates for ground motions at a particular location. On the other hand it does not seem quite adequate to define those that are

such candidates, let alone the probabilities of their occurrence.

It has been customary in the theoretical work in the field to supply the missing

information by assumption. This is alright provided the results that are obtained in this

fashion are "robust": the effect on them of variations in the assumptions must be weak.

In reliability problems in general, however, results often are far from robust, and the

problem of earthquake resistance appears to be no exception. One might go so far as to

say that any information that can safely be supplied by assumption may be unnecessary in

the first place.

134
Evidence of an analytical sort, and pointing to non-robustness, was presented in a

predecessor paper [1]. It may be of interest to add here some numerical support. It bears

more specifically on the non-robustness of assessments of structural earthquake resistance,

relative to an almost universal assumption namely, that the ground motion during an earthquake is a sample function from a Gaussian random process. The question then is whether

such assessments are affected by small variations away from the Gaussian probability

measure. The answer is that they are affected, and often quite strongly so.

Figure 1 shows two pairs of probability densities. One is Gaussian. The other is

Gaussian as well but only up to the points y 0 = ± 1.5 o. Beyond y Q is non-Gaussian and of

a form already used in [1], namely

P(y) = -

s ys exp( 1

r

r

)

The parameter a is so chosen that the transitions at y 0 are continuous. The difference

between the two densities are evidently small, as Figure 1 shows, no doubt too small to be

detected from statistical analysis of ground motion records.

The effect of the difference can be quite large however. If the exceedance of the

danger level L in Figure 1 is the assessment criterion, one finds [3] that the probabilities of such an event are

m = 0, r = 2 (Gaussian): p = 0.020

m = 2, r = 2: p = 0.352

m = 0, r = 3: p = 0.0005

The assumption of a Gaussian probability measure in other words could easily have led a

designer to underestimate the probability of trouble by a factor of about 17 or to overestimate it by a factor of about 40 ("easily" in the sense that he would have no way of

suspecting that he had made either error).

CRITICAL EXCITATIONS

The moral of this example is that a certain caution is indicated in the assessment of

structural reliability in general and of earthquake resistance in particular. By these

indications information, especially of a probabilistic nature, should not be used unless

135
it is well documented, or else its effect on the results is demonstrably weak. The

question then is how to arrive at assessments in the face of incomplete information.

The approach described in this paper is based on the idea of utilizing only

information which is documented on an adequate confidence level. This information will

then define a class of excitations which can realistically be expected to occur and which

the structure accordingly should be able to sustain. In order to assess the reliability

of the structure it is thus necessary to find those excitations in the class which induce

in it the highest stresses or strains. These are then its "critical excitations." If a

structure survives its critical excitations with the desired level of integrity, it will do

so also with all other realistic ones. Its reliability assessment can thus be justifiably

positive

.

In the context of earthquake engineering, this paper more specifically considers

information well enough documented if it contributes to the distinction between excitations

which are realistic candidates of possible strong ground motions at a given site, and those

that are not. In other words it considers only the information to be admissible which is

contained in items (a) to (f) of Section 2, after having been suitably quantified. The

quantification is arrived at by the following line of reasoning.

It starts with the observation that any ground motions already recorded at or near the

location of interest certainly are realistic. The same can presumably be said of those

recorded at locations with similar geological and seismological properties. Such records

are then called the "basis excitations" for the site under consideration. In order to

extrapolate from them to other not yet recorded excitations, all excitations are admitted

as well which are linear combinations of the basis excitations. One notes that these

combinations do share with the basis excitations many of the characteristics (a) to (f)

of Section 2. They have in common with them the durations, the intensity variations with

time, the amplitude spectra, and the seismological characteristics. To this extent therefore they represent current experience. However, the bounds on intensity (in whatever way

the term is defined), on acceleration and on velocity must be explictly imposed.

136
The problem of assessing the earthquake resistance of a structure is thus reduced to

that of determining its critical excitations, i.e. that of finding the linear combinations

of the basis excitations which, subject to the bounds on ground motion intensity, acceleration, and velocity, achieve the largest structural response peaks.

DETERMINATION OF CRITICAL EXCITATION

The problem of determining the critical excitations of a structure is the following.

It is desired to find an excitation of the form

x(t) = Sextet) (1)

in which the sum ranges over the basis excitations x^(t). It is to be so determined that

the response

y(t) = E iaiyi (t) (2)

generated by it in a designated structural variable achieves the largest peak maxt y(t)

provided that the two constraints

maXj. x(t) < M-^, max tx(t) < M2 (3)

are satisfied, as well as a third one on the "intensity" of x(t). The term "intensity"

is not uniquely defined. In this study it was interpreted (for sake of traditions rather

than of necessity) to mean the square-integral of the ground acceleration. The third

constraint thus is of the form

00/ x

2 (t) dt < M

Q

2 (4)

0

In the mathematical terminology, one would say that it is desired to find an element

x, lying in the intersection of the linear manifold spanned by the basis excitations x^

with the cubes (3) and the solid ball (4), and having the property that it maximizes the

peak of y(t) in (2)

.

This is an optimization problem. There are several ways in which it can be solved.

They differ chiefly in how much of the work is done analytically before numerical

137
algorithms are resorted to. The results reported below were obtained (again, for sake

of tradition rather than for any other reason) by a method that starts analytically.

It assumes that the time t* at which y(t) reaches its peak is known and that the problem

is that of determining the coefficients which achieve

y(t*) = maxa E^y^t*) (5)

where a stands for the vector of coefficients a^. This problem is solved subject to the

constraint (4), only. It is a least-squares problem of a rather conventional sort, with

the solution

a* = a*(t*) = ± X X"1 y (t*) (6a)

in which X is Gramian matrix of the accelerations x^(t), i.e., the matrix with the elements

00 x

ii =

/ x

i (t)x,(t) dt (6b)

0

and y(t*) is the vector with the components y^(t*). The factor A is the Lagrange

multiplier

\ = X(t*) = M

0

[y T (t*)X_1 y(t*)] 1/2 (6c)

in which yT denotes the transpose of y.

This is only a partial solution, of course. For one, contrary to what has been

assumed, t* is not known and for another, the constraints (3) are not observed. The determination of t* and the imposition of the constraints however can be carried out only by

numerical computation. To do so, one must shorten the time interval from (O, 00 ) to (0,T)

with an appropriate T, and one must discretize the time within it to a set of N equally

spaced instants t^, k = 1,2.. N, with an appropriate N. The choice of T and N is not

entirely trivial [4] but, since it is not really pertinent here, its discussion will be

omitted.

In the numerical work, one can begin by deriving the preliminary solution (6) for

each instant t^, and then use it as a starting point for a nonlinear programming routine.

This routine would seek the set a of coefficients which maximize the response y(t^) for

1 38
that instant and which also satisfies the constraints (3). The desired critical excitation

is then the one which produces the largest among the values y(t^) determined in this

fashion.

COMPUTATIONAL PRACTICE

In practice, the computational burden of executing a nonlinear programming routine at

every one of several hundred time instants t k is nearly prohibitive. To reduce it, a simpler procedure was used in the application to be described here, even though it entailed

the risk of settling for a response peak that is only a local (rather than a global) maximum. Under this procedure, the maximum peak (i.e. the maximum y(t*) among the values

y(tk)) was first found subject to the constraint (4) only and then the constraints (3)

were imposed on y(t*).

The nonlinear program routine by which this imposition was executed was the so-called

penalty function method, due originally to Fiacco and McCormick [5] and later converted to

a generally available computer routine by Mylander et al. [6], under the designation SUMT.

In order to exploit this routine, the 2N inequality constraints (3) were first combined

into one equality constant, namely

h = h(ct) = \ P (x 2 (t k

) - M

2

) + l k P (x 2 (t k) - M

2

) = 0

p(z) = z2 /2 for z > 0, and p(z) = 0 f or z < 0

.

The constraint (4) was left in its original form, i.e., as the inequality

g = g(a) = £ k

x

2 (t R

) " M

2 < 0.

Fiacco and McCormick have shown that the constrained programming problem under study can

be replaced with an unconstrained one. Its objective function is then

f = f(a,r) = y(t*) - r log g + _L h

2 r

and its solution approaches the desired one as r+°°.

Limited computational experience as of this writing indicates that, in many cases, the

nonlinear programming routine is not called at all, indicating that the constraints (3) are

139
subsumed under (4). In those instances in which SUMT was activated, SUMT performed

reasonably well (but no better than that), perhaps justifying its reputation for slow

convergence

.

COMPUTATIONAL RESULTS

A small sample taken from the body of results that were obtained with this method may

give an indication of its performance. The results were derived for the Sequoyah nuclear

reactor which is under construction at Daisy, Tennessee. An eleven-node model of the plant

was available, and so were twelve ground motion records which had been obtained at locations with geophysical properties similar to that of the Sequoyah plant. Six of these were

from the U.S.A. and six from Italy. They represented the basis excitations.

The bounds on peak ground acceleration, peak ground velocity, and intensity were set

-1 -2 3/2 at = 0.025 cm sec , M2 = 0.346 cm sec , Mq = 0.314 cm sec . The response peak generated in the displacement of node 1 by its critical excitation was found to be 3.646 cm.

By way of comparison, the largest response peak generated by any one of the basis excitations was 2.432. The former is thus higher by a factor of 1.5 than the latter.

A typical critical excitation is plotted in Figure 2.

DISCUSSION

The critical excitation method extrapolates from existing experience, regarding strong

motion at a particular location, and seeks to encompass all excitations that can realistically be expected there. It then assesses the earthquake resistance of a structure by the

largest response peak generated among its variables by these excitations.

The assessments can thus be said to be based on the largest realistic response peak.

A positive one is then a guarantee of adequate resistance, or at least as close to a guarantee as may be possible at present. In the example in the preceding section, for

instance, it is found that there are realistic response peaks that are half again in high

as those that can be derived from existing records. A comparison of the responses calculated by the critical excitation method with those obtained from the conventional response

spectrum technique shows that the critical excitation method predicts somewhat higher

response consistently.

140
Contrary to much current custom, no probability figures are attached to the results.

Such figures, on the basis of the discussion in the second major section, cannot be relied

upon and their use would degrade the confidence level of results derived by this method.

In particular an argument, according to which response peaks as high as those obtained by

this method are "highly unlikely," could not be pursued without a loss in the reliability

of the results.

This reliability is probably the main advantage of the method over others. Its main

disadvantage may be the fact that, at least in principle, a single structure has many

critical excitations (one for each of its structural variables). The determination of all

is clearly not practical. Computational experience however indicates that actually all

need not be determined (the critical excitations for many variables are so similar that

any one of them is an adequate representative of the others). However, not enough is

known in either theory or practice regarding how broadly valid such representations are

and how best to establish the range of their validity.

ACKNOWLEDGMENT

The work described in this paper was done partly with the support of the U.S. Nuclear

Regulatory Commission under Contract No. NRC-04-77-097 and partly by the U.S. National

Science Foundation under Grant No. ENV 76 - 14893. The support is much appreciated.

REFERENCES

[1] Drenick, R. F., "On a Class of Non-Robust Problems in Stochastic Dynamics,"

Stochastic Problems in Dynamics, B. L. Clarkson, ed . London: Pitman, 1977.

[2] Vanmarcke, E. H., "Structural Response to Earthquakes," Seismic Risk and Engineering

Decision, C. Lomnitz and E. Rosenblueth, ed .' s New York: American Elsevier, 1976.

[3] Drenick, R. F. and Yun, C. B., Journal of the Structural Division, American Society

of Civil Engineers 105, 1879-1891. Reliability of Seismic Resistance Predictions,

1979 .

[4] Wang, W. Y. et al .

, Nuclear Engineering and Design 52 165-174. Effective Duration

of Seismic Acceleration and Occurrence of Maximum Responses, 1979.

[5] Fiacco , A. V., and McCormick, G. P., 1968 Nonlinear Sequential Unconstrained

Minimization Techniques, New York: Wiley, 1968.

[6] Mylander, W. C. et al., "A Guide to SUMT - Version 4," U.S. National Technical

Information Service No. Ad-731-791, 1971.

141
:CAUSSIA\ :NON Gaussian (»-2. r - 2)

Fig. 1. Probability Density Functions

1 1 1 1 1 1 —

1

00 2.00 4.00 6.00 8.00 10.00 12.00 14.00

TIME SEC

Fig. 2. Example of Critical Excitation

142
DYNAMIC TESTING OF A THREE- STORY REINFORCED CONCRETE STRUCTURE

by

Keiichi Ohtani

ABSTRACT

Dynamic tests using large-scale earthquake simulators are useful for confirmation

of aseismic properties of reinforced concrete structures. This study presents a summary

of earthquake simulator tests of a reinforced concrete structure at the Earthquake Engineering Laboratory of the National Research Center for Disaster Prevention, located at

Tsukuba New Science City.

A full-size, three-story, single-bay reinforced concrete frame was built on the

shaking table. The floor area of the frame was 100 m^, the total height 10.95 m, and

total weight about 285 tons. Two sets of experiments were performed. In set (Series I),

only the frame was actuated, and in the other (Series II) both the frame and non-structural

elements which were attached to it were actuated. The experiments were carried out in several stages of simulated and modified earthquake ground motion with an intensity sufficient

to cause inelastic behavior and dynamic property changes in the frame. The dynamic behavior of the frame and the influence of the non-structural elements were investigated. In

both Series I and II experiments, the change in natural frequency and the change of dynamic

properties in the plastic elements were examined.

This study is a part of the U.S. -Japan Joint Earthquake Research Program.

143
INTRODUCTION

The best way to examine the aseismic properties of a structure is the dynamic test

using the actual size structure. However, in experimental studies, test specimens are

element tests or reduced scale models. It can also be noted that most tests of actual

size structures are performed only in the static condition. Fortunately, the large-scale

shaking table of National Research Center for Disaster Prevention has a capacity large

enough to excite a model structure of up to 500 tons. Therefore, these investigations

have been performed using one story, single bay reinforced concrete structure in actual

size. [1]

After the 1968 Tokachi-oki Earthquake, both Japanese and American earthquake engineers

saw the necessity of cooperative research projects. On August 10, 1979, the Implementing

Arrangement of the U.S. -Japan Joint Earthquake Research Program was signed by the Science

and Technology Agency, Ministry of Construction for Japan and the National Science Foundation for the U.S.A. One objective is to improve the aseismic properties of reinforced

concrete structures. The Japanese National Research Center for Disaster Prevention has the

task of producing the basic data for the pseudo-dynamic test of a seven-story reinforced

concrete structure in actual size by performing an earthquake simulator test of a threestory reinforced concrete structure in actual size.

The results of experiments with a reinforced concrete structure as one of the

preliminary tests of the U.S. -Japan Joint Research Program are reported here.

OUTLINE OF TEST SPECIMEN

The frame of three-story, single bay reinforced concrete was examined in this test.

The test was carried in the following two cases.

(1) The frame alone (Series I).

(2) The frame with a non-structural element, a partition wall of gypsum board between

the frame in the excitation direction (Series II).

Figure 1 shows the general view of test frame. Figures 2 and 3 show the setting plan

of the foundation and the typical floor. Figure 4 shows the section of the excitation

direction. The total height of this frame is 10.95 m, foundation girder height is 1.2 m.

Story height and column height of the first floor are 3.75 m and 3.25 m, respectively.

144
Story height and column height of the second and third floors are 3.0 m nd 2.5 m. Frame

widths are 6 m in both directions. The floor slab overhangs 2 m from the center of the

column section in each direction. Each column section is 50 cm x 50 cm; floor girders in

both directions are 50 cm high, 30 cm wide; and each floor beam in the tangential direction

of the excitation is 45 cm x 25 cm in section. Slab thickness is 12 cm on each floor.

(These sectional properties are common to each floor.) Each column footing is 3 m x 3 m in

plan, 60 cm in height. The footing was fastened to the shaking table by four 2 inch steel

bolts (SCM-3) at 3 or 4 positions.

This frame was designed and constructed in accordance with the Calculation Code for

Reinforced Concrete Structures issued by Architectural Institute of Japan. The reinforcing

steel bars used in the frame were deformed bars, SD35, having a nominal yielding stress of

3.5 t/cm2 . Concrete was a normal one with design strength of 270 kg/cm2

. However, as

there was not enough time for the third floor concrete to cure after placement, high-earlystrength cement was used on that level. Table 1 shows the list of the sectional properties.

A concrete weight of 2mx4mx0.5m size is set at two places on each floor to

simulate a live load. Therefore, the slab of each floor is made in consideration of a live

load of 80 kg/m2 . The total weight of the frame is about 235 tons, including foundation

parts. For the convenience of crack observations the surface of the frame was coated with

white water paint and grid lines at intervals of 30 cm.

The experiment began about three weeks after the final concrete placement. One day

before the experiment, the tested compressive strength of concrete was 260 kg/cm 2 in average value for the final concrete placement. The other concrete placements were over design

strength. (Concrete was placed at four separate times.)

Series II experiments started two weeks after the Series I experiments. Repair and

retrofitting work took place during this two week interval. Clear cracks which occurred in

the Series I experiments were filled with epoxy resins. After they hardened, we equipped

the excitation direction frame (A, B frame) with the partition wall non-structural elements. These partition walls were made of 15 mm thick gypsum board and light gauge steel

attached to columns and girders. The studs were set at intervals of 90 cm with (45 x 40 x

10 x 0.6] steel. The gypsum board was fastened to both surfaces of the studs with damping

145
screws. We used this method to erect three walls on the first floor and two walls on the

second and third floors. These are walls which resist fire up to three hours on the first

floor and two hours on the second and third floors.

EXPERIMENTAL APPARATUS AND MEASURING EQUIPMENT

The shaking table can move in one horizontal direction by electro-hydraulic drive.

It is 15 m x 15 m in size, 160 tons in weight. The maximum attainable displacement and

velocity is ± 3 cm and 37 cm/ sec . The table may be used on specimens weighing up to

500 tons, and may be excited to 0.55 G in acceleration with full load. The natural

frequency of the system is about 20 Hz, and the operational frequency band is 0-50 Hz.

The dynamic behavior of the test frame and table can be measured in several ways:

Acceleration of the table and each floor by strain gauge type acceleration pick-ups,

relative displacement of each floor with ring displacement transducers, table displacement with IDS type displacement transducer located on the shaking table foundation, and

wire strain gauges on the main reinforcing steel bars of each column. In Series II, we

measured the strain on the gypsum board and the relative displacements between the gypsum

and the concrete frame. The test data were recorded using a 64 CH digital data recorder;

main data were monitored with a 14 CH analogue data recorder, and two CH pen recorders.

We recorded the condition of cracks at the bottom of column on the first floor with a

video recorder.

TEST SCHEDULE

Table 2 shows the test schedule. The STEP vibration uses the rectangular wave, a

vibration amplitude of PP-1 mm, and a period time of 10 sec or 20 sec for the free vibration of the frame.

The SMAC-M record of the N-S component of the 1978 Off-Miyagi Pref . earthquake

recorded at Tohoku University installed on the first floor was the table input for the

earthquake motion test. The earthquake record was digitalized by the Building Research

Institute, Ministry of Construction.

When the earthquake ground motion is simulated, because of the limitation of the

3 cm shaking table stroke, the low frequency components of the original earthquake motions

146
may be cut off in order to get high intensity shakes. Therefore, table inputs for the

earthquake motion tests were obtained as outputs from the one-degree-of-f reedom system

response to the original earthquake motions. The damping constant and the natural f remodification to the original earthquake motion was achieved using a 2 Hz high-pass filter

in the acceleration. With this modification, we were able to vibrate up to about 400 gal

within the 3 cm stroke limit.

Table 3 shows the maximum acceleration responses and maximum displacements obtained.

The maximum acceleration responses of Table 3 are values on the shaking table (AT) , the

second floor level (A2), the third floor level (A3) and the roof floor level (AR). The

maximum displacements are defined as follows: DT is the relative displacement between the

shaking table and its foundation; and the displacements at the first floor (Dl), the second

floor (D2) and the third floor (D3) are the story displacements between each floor levels.

The story deflection angles between each floor level are also given in Table 3.

Test result from each series are given below.

This is the testing of the reinforced concrete frame. By microtremor measuremeut

before the tests, the natural frequency of the frame was determined to be 2.8 Hz.

The first run of the simulated earthquake ground motion was set to produce an

acceleration of 0.03 g on the shaking table. From this, a maximum acceleration on the

shaking table of 36 gal and an acceleration response on the roof floor of 90 gal resulted.

The story deflection angles on each floor were 1/1100 - 1/2000. Cracks were not observed

on the columns and the girders. The natural frequency of the frame was 2.22 Hz after the

test. An extension of the natural period of about 30 percent was observed when compared

to the microtremor measurement.

The second simulated earthquake ground motion run was aimed at an acceleration of

0.15 g on the shaking table. The resultant accelerations on the shaking table and the

roof floor were 140 gal and 370 gal respectively. The story deflection angle on each

quency of this system is taken as In other words, this

TEST RESULTS

Series I

147
floor was 1/220 - 1/480. At the bottom of column of the first floor, cracks were observed.

After the test, the natural frequency of the frame was 1.28 Hz and the extension of the

natural period reached to 70 percent compared to the first run result.

The third simulated earthquake ground motion test was set to produce an acceleration

of 0.25 g on the shaking table. The accelerations on the shaking table and the roof floor

were 220 gal and 430 gal respectively and the story deflection angle of each floor was

1/120 - 1/300. Figure 5 shows the time history. Cracks were observed at the top of column

at the first floor, the girders at the second floor and even at the top of column of the

third floor. Following this test, the natural frequency of the frame was found to be

1.18 Hz and an extension of the natural period to 88 percent compared to the first run

result

.

The fourth run of the simulated earthquake ground motion was set to produce an

acceleration of 0.4 g on the shaking table. The accelerations on the shaking table were

320 gal, or 80 percent of the initial target. However, the displacement of the shaking

table was 29 mm, with 30 mm the specified performance. This means that the shaking table

performs as it was designed. The acceleration response on the roof floor was 500 gal and

the story deflection angle of each floor reached 1/110 - 1/280. The ring type displacement

transducer at the second floor that measures story deflection angles up to 1/90 was broken

at its mounting part. Therefore, the actual value of the story deflection angle of the

second floor could not be measured, but is assumed as almost 1/100. After the fourth run,

the natural frequency was 1.14 Hz which is nearly equal to the result of the third run.

From this, it could be said that the dynamic properties of the frame scarcely changed

after the third run. Figures 6 and 7 show the crack behavior after the Series I tests.

They show portions of the bottom of column at the first floor and the top of column of the

third floor. These cracks were not serious and the members were not pushed to maximum

strength. The frame shown its durability under a condition of 0.4 g earthquake motion

and the design criteria of this frame is justified by these tests.

Series II

In this series, the non-structural element attached to the frame was tested. Before

this vibration test, the microtremor of this frame was observed and the natural frequency

estimated at 3.3 Hz. From the retrofitting of epoxy resins and the rigidity of the gypsum

148
board elements, the natural frequency of this measurement is higher than that of Series I.

The damping constant of this measurement was estimated at four percent. (In the Series I

tests, the damping constant was estimated at two percent.)

In the first run of the simulated earthquake ground motion, in which 0.03 g was the

target acceleration at the shaking table, we obtained a 27 gal maximum acceleration on the

shaking table and a 65 gal acceleration response on the roof floor. The story deflection

angle of each floor was 1/2000 - 1/4500. These values were almost half those of Series I

results. We estimated that the gypsum board exercised a control capacity on the displacement at the small level excitation.

From the second run in which 0.25 g was the target acceleration of the shaking table,

we obtained 240 gal of the table acceleration and 350 gal of the roof acceleration.

Figure 8 shows the time history. During this run, the light gauge steel, which was holding

the gypsum board, was observed to move. The reason is that the carl plug, which affixed

the light gauge steel to the lower surface of the girder, fell off during the acceleration.

Though concrete nails are used to affix the column surfaces on the upper surfaces of the

girders, a carl plug is a generally used techique on the lower surfaces of the girders.

This experiment taught the following: a carl plug can resist a fairly big static load,

but it is weak under a dynamic load. Then the measure to fix the girder surfaces needs

important. The story deflection angle of each floor was 1/180 - 1/330 and these values

increased, which proves that the gypsum boards do not contribute to reinforcement expected

and that the effect of the reinforcement was lessened by the dynamic load.

In the third simulated earthquake ground motion trial in which 0.4 g was the target

acceleration of the shaking table, obtained a 370 gal table acceleration and a 500 gal

roof acceleration. Figure 9 shows the time history. The story deflection angles of each

floor ranged from 1/130 to 1/310, and these values are approximately the same as those in

the Series I frame only test. The change of the natural frequency was as follows: The

natural frequency was 1.85 Hz after the first run, and the natural period increased by 80

percent compared to what it was before the experiment. The natural frequency was 1.22 Hz

after the second run, and 1.19 Hz after the third. These values are approximately equal to

those of Series I. In this experiment, the crack, which had been repaired in Series I,

appeared again somewhat enlarged. However, no new cracks were observed.

149
The arrangement and analysis of the data are now under way. When it has been

completed, a more detailed report will be made.

REFERENCES

[1] "Dynamic Behavior of Reinforced Concrete Frame Structures," by K. Ohtani, and

C. Minowa, 11th Joint Meeting of U.S. -Japan Panel on Wind and Seismic Effects,

UJNR, September, 1979.

150
Table 1. List of Sectional Properties

Portion Section

(cm) Arrangement of Reinforcement

Column 50x50 8-D22 Hoop; D10-§100

Girder 50x30

Center J

0

?' \'TML Bottom; 2-D22 _ .

.

Stirrup;

Tori • A—vtoo m n at or> Bottom; 2-D22

Beam 45x25

Tod: 2-D22 Center ^' Bottom; 2-D22 Stirrup;

, Top; 3-D22 D10-@100

Bottom; 2-D22

Slab 12 Top; Dl3-@200 Mesh in

Bottom; Dl3-@200 both axis

Footing G. 120x50

Bottom; 3-^25 Stirrup; D16-@200

Table 2. Test Schedule

Series Test No. Input Condition

STEP1 Rectangular PP-lmm 0. 1Hz

RUN1 Earthquake 0.03G

RUN2 Earthquake 0.15G

I STEP2 Rectangular PP-lmm 0. 1Hz

RUN3 Earthquake 0.25G

STEP3 Rectangular PP-lmm 0. 1Hz

Run 4 Earthquake 0.4G

STEP 4 Rectangular PP-lmm 0. 1Hz

STEP1 Rectangular PP-lmm 0. 05Hz

RUN1 Earthquake 0.03G

STEP 2 Rectangular PP-lmm 0. 1Hz

n RUN2 Earthquake 0.25G

STEP3 Rectangular PP-lmm 0. 1Hz

RUN 3 Earthquake 0.4G

STEP4 Rectangular PP-lmm 0. 1Hz

151
Table 3. Test Results

Series Test No. Maximum Acceleration (gal) AT A2 A3 AR

Maximum Displacement (mm) (Story Deflection Angle) DT Dl D2 D3

I

RUN1 36.3 61. 3 68. 3 88.6 2.34 2.77

(1/1173) 2.19

(1/1142) 1.26

(1/1984) RUN2 143.1 206.7 301.1 370.2 13.09 14.88

(1/218) 11.09

(1/225) 5.21

(1/480) RUN 3 217. 5 233. 3 363.9 427.1 20. 86 25.53

(1/127) 17.84

(1/140) 8.49

(1/294) RUN4 322.4 358.8 396.5 498. 3 29.09 27.99

(1/116) 8.92

91/280)

n

RUN1 27. 5 39.4 58.7 65.1 2.51 1.45

(1/2241) 1.20

(1/2083) 0.55

(1/4545) RUN 2 235.6 259.3 359.0 374.5 19.07 17.32

(1/188) 14.10

(1/177)

7. 50 (1/333) RUN 3 369.9 432.1 473.0 498.0 28.90 25.40

(1/128) 16.24

(1/154) 7.96

(1/314)

Fig. 1 General View of the Test Frame

'5000

6000

.f'i.

F FG 1 C

3000 30QQ

1200Q

' SHAKING TABLE

=£> EXCITATION DIRECTION

Fig. 2 Setting Plan of

the Foundation

152
© ©

®

®

i=3I

:

c

2000 =li

G

C G

3

1

s 0009

1

2000

2000 | 6000 1 2000 |

Fig. 3 Setting Plan of Figthe Typical Floor

2000 6003 2000

1 1 Tf 1 -I

3F r~~" 1 ! " If

2F B 1

TABIF ill I I ® I

4 Section of the Excitation Direction

153
B3.35CAL

501.S7GAL

3B_28f*f

01 1

.

' " vvv>w\J^w^/'/WvW^Mr a/> - . ,=TMww~ —*

vs^ww\Z\a^A/WV\A/\A/

rt-^-r'^r r:

.

: . : . : .-

.

. . / .

Fig. 5 Time History for the RUN 3 of Series I

All

432J2GAL A 2 I .

.485.14 GAL

A R

321 14 GAL

I776MM

D2 | -vaaww\/VvW\/\/V\AAAAAAa/wv^— ^—

Fig. 8 Time History for the RUN2 of Series H

154
_413.13 GAL AT

_8642^GAL

A2

_1212.85&AL A3

_821KO\L

-4

_ 2134 MM DT

-w^/\AtW\M/>A/W\AAaa/

_ 3552 MM

D1

,_3570MM

D2

J205MM

D3

Iggtj,

Fig. 9 Time History for the RUN 3 of Series H

155
BUCKLING ANALYSIS OF BURIED PIPEINES UNDER SEISMIC LOADS

by

C. C. Chen

T. Ariman

L. H. N. Lee

ABSTRACT

The Donnell and Flugge forms of the stability equations of cylindrical shells are

employed to analyze the axisymmetric , elastic quasi-static buckling of buried pipelines

subject to seismic excitations. Using shell dimensions and the stiffness of the soil

medium surrounding the pipe as parameters, a series of numerical results are obtained,

which shows that no significant half sine wave occurring in the circumferential direction

for relatively long pipes when axial buckling load reaches its minimum. It also is shown

that for a given pipeline an increase in soil stiffness causes a decrease in wavelength of

the critical-load instability mode and, consequently, causes an increase in axial critical

load

.

156
INTRODUCTION

Buckling failure of long buried pipelines during a severe seismic excitation has

been recognized as an important problem in the pipeline industry. It has been recently

attracted to the attention of earthquake engineers and researchers i.e., [1]« These studies, however, have concentrated on a long circular cylinder or tube surrounded by uniform

soil medium, and subjected to uniform, static, radial pressure. Studies on the critical

buckling load of buried pipelines due to axial compressive loading are rarely found in

the literature. To the best of the authors' knowledge, the study on this subject was

usually simplified and restricted to the model of a beam on an elastic foundation [2]

.

The purpose of this work is to investigate the buckling failure mode of buried

pipelines under seismic excitations, using a model of a cylindrical shell surrounded by

uniform springs. In the analysis, Donnell shell equations are used for quasi-shallow

shells, and Flugge shell equations are employed when the shallow shell equations are not

applicable. Attention is focused on the parametric studies that concern the effects on

dimensions of the pipe itself as well as the stiffness of the soil medium surrounding the

pipe.

FORMULATION

Consider a thin-walled circular cylindrical shell of finite length L, wall thickness

h and mean radius a, with h«a, surrounded by a uniform soil medium with extensional stiffness k, in the radial direction. On the middle surface of the cylinder, the cylindrical

coordinate system x and 8 is placed. Distances from the middle surface are measured by a

coordinate z, positive outward. Displacement components in the x, 8 and z directions are

denoted by u, v and w, respectively. The cylindrical shell is considered to be subjected

to edge loading and to static axial compressive loading P.

Donnell' s Form of the Stability Equations

Using the stationary potential energy criterion, the nonlinear equilibrium equations

of the Donnell 's form are written as follows [4].

157
a Nx,x + Nx6,9 = 0

a Nxe>x + N 9>Q = 0 (1)

DV 4„ + lN9 -(Nx

w

>xx +i-Nxe W)XQ +^ N

0 w>ee

0

where Nx and Nq are in-plane normal force intensities, Nxg is in-plane shear force

intensity, V is the Laplace operator and ( ) = — ( ). The linear equilibrium equa-

' X

i 3xi

tions can easily be obtained by the omission of all quadratic and high-order terms in u, v,

and w from Eq. (1). Introduction of the constitutive and kinematic relations to these

linear equilibrium equations yields

a2u

, xx + u)u ,ee + p)a v

,x6 + a^ w

, x = 0

-kl + p)aU)X9 +1.(1 - U)a 2v

>xx + v

)0e + w

)9 = 0 (2)

DV 4w + J_ C(v fi + w + auu v ) + kw + Pw vv = 0

a

z

In Eq. (2), y is the Poisson's ratio and C and D are termed extensional and bending

stiffness parameters, respectively, and are given by

C = Eh

1 - p2

and (3)

D = Eh3

12(1 - u 2

)

in which E is Young's modulus.

With the aid of linear equilibrium equations given in Eq. (2), the linear stability

equation of the Donnell's form can be obtained by either the adjacent-equilibrium or the

minimum potential energy criterion and written in the form

158
a2u*,xx " ^u*,66 + f <* + ^> a v*,x6 + a ^.x =

0

iCl+y)au* >x0 +-i(l- y)a2v* jXX + v*

>ee +w* >e = 0

DV 4w* + -L C(v* g + w* + ay u* ) - (N*w* +JL N* Qw*x6

a

2 ' a

+ J_N*w* eQ ) + kw* + Pw* )XX = 0 a

z

In Eq. (4), u* v* , and w* are the displacements of the buckled shell and Nx , Ng, and

Nxq are governed by the linear form of Eqs. (1) and can be obtained for a given edge

loading condition.

It is to be noted that Eqs. (4) represent a coupled set of three equations in the

variables u* , v* , w*. These equations may also be partially uncoupled and written as

follows

.

V

4u* = -y v u — w* + ^ w* TM

w ,xxx ^—* ^ ,xQQ

a aJ

V

4V* = -2 + V v • w* „-i w* * nn —5— w

,xx6 —r w

,eee

a

z a

H

DV 8w* + Ll-f- Cw* )XXxx - V

4 (N

xw*>xx +iNx6 w* >x6

a

z a

+ _L NflW* flfl ) + kV 4w* + PV 4w* „ = 0

Flugge's Form of the Stability Equations

In a sense, the Flugge's stability equations can be regarded as Donnell's stability

equations plus a "modifying" part.

The Donnell's stability equations given in Eqs. (4) in the absence of extensional

stiffness of soil medium k and axial compressive force P can be written in a matrix form

as follows.

[LD] {x*} = {0}

where [Lp] and {x*} are a matrix differential operator and a displacement vector

respectively and are given by

159
[Id]

.2 3' a~ ' +

3x2 2

Sym.

1-U 8'36 2 L+u a 8^

2 3x36

1-H a

2

3'

ua

+

92

30 2

1 +

a2h2 V

4 12

3x

336 (7)

and

(8)

The matrix differential operator according to Flugge's theory can be obtained and

treated as the sum of two operators,

[LF ] = [LD ] + g [LM0D ] (9)

where [L^odI ^ s a "modifying" operator and £ is the nondimensional shell parameter defined

by

12a 2 (10)

The modifying operator in Eq . (9) takes the form

[ lmodJ

1-U 3^

2 99 2

Sym.

3(l-u)

a

2 3'

3x<

- a

3 3 3 + ^ a

33 3x3 2 3x3e 2

3-p

a

2

3' 3x2 39

1 + 2 39 2

(ID

Including both the soil medium effects plus the axial compressive load in Eq. (9), the

matrix differential operator given in Eq. (7) becomes [3]

160
[LD] =

(l-P)a 2 9 +

C 3x2

Sym.

1-U 3^

2 362

1+P 3x38 3x

C 1"" -I)* 2 9

'

3x'

(1 a k) + I a

2 3

3 9'

2

3x2

330

a

2h

2 12

(12)

APPLICATIONS

In the following, parametric studies on the critical load analysis for a long simply

supported cylindrical shell are considered. Since the axial buckling is the main concern

in this study, only the seismically induced axial load is considered in the analysis.

Donnell's Axial Critical Load

In the absence of edge loading, the stability equation governing the deformation in

the radial direction becomes

DV 8w* +

I-V' ^.xxxx + kvV + PvV ,xx =

0 (13)

Solutions of the form

w* = A sin mx • sin n8 (14)

where A is an arbitrary constant, m = nnr/L, and m,n = 1,2,3..., are seen to satisfy

both the differential equation and the assumed simply supported boundary conditions. It

is noted that the quantities m and n represent the number of half sine waves in the axial

and circumferential directions respectively.

Substitution of Eq. (14) into Eq. (13) gives

P = k- [k + D(m 2 +JL

2

) 2

] + • ?L

m2 a2 a2 ,2 , n2,2 (m + 2>

(15)

A distinct eigenvalue P corresponds to each element of the set ({m}x{n}). A preliminary

study in this work, has shown that, when the axial compressive load P reaches its minimum,

161
the cross section remains circular. It is therefore reasonable to assume that the shell

deforms in such a way that the shape is totally governed by the number of axial half sine

waves. Assuming

w* = A sin (16)

and substituting Eq. (16) into (13), the counterpart of Eq. (15) becomes

P = D

2 2 i 2 T 2 m * + (k + klL C) L

2 (17)

The axial critical load P cr can then be obtained by minimizing Eq. (17) with respect to m

as follows:

Eh' cr1 + a 3(l-u 2 >

1/2

(18)

where a is the nondimensional parameter,

(Eh/a2

)

(19)

The number of half sine waves in the axial direction is

L 12(l-tQ(l+q)

11 ' a

2 h

2

1/4

(20)

The Batdorf parameter Z [4] is given by

T 2 , 1/2

ah

(21)

Then, substitution of Eqs. (3) and (21) into Eqs. (17) and (20) gives

1

n, 2 + 12(l+a) . Z_, . 2/2 1 cr

7r m

(22)

and

162
m = Z [i2(L+o)] i/ 2

TT (23)

Figures 1 and 2 show the variation of critical axial load, P cr , as a function of the

Batdorf parameter Z for various values of m, the number of half sine waves. In Figure 1 a

is zero while in Figure 2 a is unity. Similar curves for other relevant values of a would

be obtained and utilized in the determination of the critical load. Therefore, this

procedure could be useful in the design of buried pipelines.

Flugge's Axial Critical Load

Due to the shallowness, limitation, in the case of moderate or short cylindrical

shells, Flugge's stability equations should be used rather than Donnells 1 .

For a simply supported shell, the displacement {x*} has the components

u* = A* sin n6 cos Xx

a

v* = B* cos n 9sin Xx (24)

a

Xx w* = C* sin n6 sin

a

where A*, B*, C* are arbitrary constants and

X = ""ra (25) L

with m being the number of half sine waves in the axial direction of the shell. Thus

Eqs. (9)-(ll) yield

A*[X 2 + J^P n 2 (l+£) - $X 2 ] + B*[ - I+t

J Xn]

2 2

+ C*[-Xu-?(X 3-ilii X n

2

)] = 0

2

A*[_l+M xn] + B*[n2 + l~V X

2 (1+35) - *X 2

] + C*[n-3±H ?X 2n] = 0 (26)

2 2 2

A*[-Xu-5(X 3- 1_ lJ Xn2

)] + B*[n + ?X 2n] + C*[l + ?(X 4 +

2 2

2X 2n

2+n 4 - 2n 2+l) + (l-p 2 )a-$X 2 = 0

where

163
$ =

P = lUzu!l (27)

C Eh

The above represents three linear equations with the buckling amplitudes A*, B* , C* as the

unknowns. For a non trivial mode, the determinant of Eq. (26) must be zero. For small

values of the parameters £ and <J>, it is sufficient to retain only linear terms. Thus the

determinant becomes,

ABC + ABc£ -ABA 2* + 3AbC£ -AA 2 C3> +aBC£ -BA 2 C$

+ 2DEF + 2DEf? + 2DeF£ -BE 2 - 2BEe? -3bE 2£ + A 2 E

2 <I> -AF 2 - 2AFf £ -aF2 £

+ A 2 F

2 <t> -DC 2 - cD2

C + A 2 D

24> = 0 (28)

where A = A 2 + a, B = n

2 + b, C = 1 + (l-u 2 )ct, D = i±E An, E = Au, F = n (29)

2

with a = llH. n

2

, b = lUd A 2, c = A 4 + 2 A 2 n

2 + n4 - 2n2 + 1 2 2

e = -A(A 2 - a), f = A 2n (30)

2

Solving for <t> in Eq. (28) yields

* = * • JL (i^)

2 (31)

S TT 2 ma

where R = (ABC + 2DEF - BE2 - AF2 - CD2

) + (ABc + 3AbC + aBC + 2DEF

+ 2DeF - 2BFe - 3bE 2 - 2AFf - aF 2 - cD2

) S,

S = AB + AC + BC - D

2 - E2 - F2 (32)

Figures 3 and 4 give the variation of $ which is related to the axial buckling load,

as a function of the non dimensional parameter L/ma. Both figures are for a ductile iron

pipe of L = 40 ft, a = 24 in, h = 0.51 in and £ =3.763 x 10~5 . They differ in that the

higher value of a in Figure 3 is 0.01 while in Figure 4, it is 0.5. Both figures also give

the case of no soil medium around the pipe (a = 0) .

The stability of the above pipe under seismically induced axial loads can be examined

using Figure 3 and 4 for selected values of L/ma (m = 1,2,3, etc.) and the corresponding

values of $ . As long as all the points for specific values of L/ma and $ lie below the

heavier dashed lines (no soil) or the solid lines (with soil) the pipe is stable. When the

164
axial load P and hence <t> increases, all the points move upward. As soon as any point

reaches one of the curves, the pipe is in neutral equilibrium and about to buckle.

Figures 3 and 4 show that the soil medium has an important effect in the critical axial

load by causing an appreciable increase in P cr .

ACKNOWLEDGMENT

This work is supported by National Science Foundation under the Grant No. ENV-77-23236

in which Dr. S. C. Liu is the Program Director.

REFERENCES

[1] Cheney, J. A., 1971, "Buckling of Soil-Surrounded Tubes," Journal of the Engineering

Mechanics Division, ASCE, Vol. 97, No. EM4 , pp. 1121-1130.

[2] Ariman, T., Muleski, G. E., 1979, "A Review of the Response of Buried Pipelines under

Seismic Excitation," Lifeline Earthquake Engineering - Buried Pipelines, Seismic Risk

and Instrumentation . Edited by T. Ariman, S. C. Liu and R. E. Nickell, ASME, PVP-34

,

pp. 1-29.

[3] Muleski, G. E . , Ariman, T., and Aumen, C. P., 1979, "A Shell Model of a Buried Pipe

in a Seismic Environment," J. Pres. Vessel Tech. ASME, Vol. 101, pp. 44-50.

[4] Brush, D. 0. and Almroth, B. 0., 1975, Buckling of Bars, Plates, and Shells , McGraw-Hill, Inc.

165
100,000

10 100 1.000

z 100,000

100,000

—, 10,000 V

7,

l '000

D 100

~> 1 1 r

o I 0 "

- m . 10 /// 30 / / / / 40 /////—50 //// / 60

. 7080

„ 90

100

//////

'1 1 1 I - -t1 10 100 1.000 100.000

Fig. 1 Nondimensional critical axial Fig. 2 Nondimensional critical

load P versus Z. load P versus Z. cr cr

166
SHAKING TABLE TESTS OF EARTH RETAINING STRUCTURES ON IMPROVED GROUND

by

Setsuo Noda

Sosuke Kitazawa

Hajime Tsuchida

ABSTRACT

In coastal area land reclamation work, bulkheads are constructed around filled-in

ground. In the case of gravity type bulkheads, the soft ground is improved. Instead of

conventional improvement methods a Deep Mixing method (D.M. method) has been used recently.

A design procedure for these structures has been tentatively proposed. It has not

been tested because the structures have yet to experience strong earthquakes. In order to

improve earthquake resistant design procedures, shaking table tests are being carried out

on five models of gravity bulkheads on improved ground.

This paper presents model outlines, test procedures, some experimental results,

stability analyses of the present design procedure and estimates of prototype stability.

167
INTRODUCTION

In land reclamation work in a coastal area, the soft ground is usually improved and

bulkheads are constructed around the filled-in ground. In using gravity bulkheads, the soft

ground is excavated and replaced with sand of sufficient bearing capacity.

Instead of conventional ground improvement methods a Deep Mixing method (D.M. method)

has been recently used in coastal work due to environmental restrictions. In the D.M.

method soft clayey ground is stabilized chemically with quick lime or cement milk which is

squeezed into the ground and mixed together. The strength characteristics of treated soil

have been investigated in the laboratory and practical construction features are tested in

the field. A design procedure for bulkheads and improved ground have been tentatively

proposed. [1,2] It has not been tested because the number of these structures is limited

and they have not yet experienced strong earthquakes.

In order to improve the earthquake resistant design procedure, shaking table tests are

being carried out on five models of gravity bulkheads on improved ground. There are two

wall types, caisson and cell; and two ground improvement types, block and pile.

This paper presents model outlines, test procedures, some experimental results,

stability analyses of the present design procedure, and the estimates of prototype

stability.

PURPOSES OF TEST

Much research has taken place in Japan to establish earthquake resistant design

procedures for the above new soil structures. In the Ministry of Transport, a program of

investigation, shown in Figure 1, is now underway. Short of a destructive test on the

prototype, it is difficult to estimate soil structure stability under severe earthquake

disturbance conditions accurately from only one method. Therefore, the earthquake resistant design procedure will be established by making effective use of the proven features

from different models.

The model tests, now in the early stages, presented in this paper correspond to the

investigations heavily outlined in Figure 1. The model tests have the following purposes:

168
an understanding of the differences in seismic behavior for the various improvements and

wall types, the collection of data to examine the numerical simulation method, and

estimates of the prototype's seismic stability.

In shaking table tests of soil structures it is often difficult to satisfy the scale

effects and similitude factors. However, tests on large scale models which are regarded as

small prototypes can be carried out since a shaking table with a large loading box, capable

of producing large exciting forces has been constructed recently. From tests of these

prototypes, differences in seismic behavior between the prototypes under varied earthquake

conditions may be estimated.

A numerical simulation method will be useful in estimating behavior of the whole

structure, but the method has not been completed yet. In order to examine the method and

apply it to practical design work, information on dynamic soil properties and on seismic

behavior is required. This information can be obtained from these model tests.

Although the models are relatively large, they are small prototypes. Therefore,

taking the similitude between model and prototype into consideration as well as possible,

the seismic stability of prototype is estimated from the experimental results for trial.

MODELS AND TEST PROCEDURES

Similitude

Since the purpose of the tests is to simultaneously obtain fundamental information

from several standpoints, the model is regarded as both a small prototype and a scaled

model. In determining model size and shape, the gravity type bulkhead on improved ground

in the coastal area was chosen as the prototype where water depth is -20 m and soft clay

extends up to -40 m.

Considering the loading box size (8m length, 4m width, and 2 m depth), 1/20 is the

adopted ratio of length (Lr). Since the two kinds of silty-clay which exist off Kawasaki

in Tokyo Bay (Kawasaki-A and Kawasaki-B clay) are used, the model unit weight is almost the

same as that of the prototype. The acceleration on the model is also the same as that on

the prototype. The ratio of cohesion of soils is 1/20, the strength of the model ground

is measured by the uniaxial compression test for treated soil, and the vane test is used

to measure the clayey ground.

169
Many variations in ground improvements are applied in practical work. For example, the

soft ground may be 100 percent treated uniformly (block type), or may be treated as group

piles (pile type). A wall type and a checkered pattern type are also used as variations in

improvement types. In this report the block type and the pile type are modeled. In the

pile type, the average strength (02) of composite ground is defined as follows:

c 2 = (1 - B) c 0 + g Cl .

cq is the cohesion of untreated soil, ci is the cohesion of treated soil, and 3 is the

ratio of the cross-sectional area of treated soil to the whole one, and is called the

improvement ratio. Because it is difficult to make many piles of treated soil of short

diameter accurately, the strength and the diameter of treated soil and the improvement

ratio do not satisfy the similitude. However, the average strength of composite ground

(c 2 ) does satisfy the above mentioned similitude. Similitude is summarized in Table 1.

3.2 Models

Model details are shown in Table 2. Models 1 to 4 are caisson type bulkheads, and

Model 5 is a cellular type bulkhead. The improved grounds are block type in Model 1 and

2, and pile type in Model 3 to Model 5. In every model, ground improvement extends to sand

of sufficient bearing capacity, and that is called fixed type.

(1) Model 1 (See Figure 2)

In this model the strength of the improved ground is larger than expected, because of

the difficulty in soil quality control and in the differences in curing conditions of

treated soil between the model and the preliminary laboratory test. Compared to the other

three caisson models, Model 1 has a wider improved ground area and a thicker gravel mound.

It is judged the most resistant to earthquakes.

(2) Model 2 (See Figure 3)

Since Model 1 was strong enough in the tests, Model 2 is designed as a rather weak

model. That is, the strength of treated soil is about 60 percent of Model 1 and the

improved ground is smaller. Additionally, the gravel mound is 15 cm thinner and the

caisson is 17 cm higher than those of Model 1.

170
(3) Model 3 (See Figure 4)

The improved ground consists of 99 piles of treated soil of 15 cm in diameter, and

the improvement ratio is 10 percent. Pile diameter is 1.5 to 3.0 times as long as required

from the similitude, but the improvement ratio is about 1/6 to 1/5 of prototype.

(4) Model 4 (See Figures 5 and 6)

Cross section and ground plane of Model 4 are shown in Figures 5 and 6, respectively.

The improvement ratio is 20 percent, and the ground contains 208 piles. The improvement

ratio of this model is about 1/2 to 1/3 of the prototype and resembles the prototype more

than Model 3. In order to equalize the average strength of the composite grounds, the

treated soil strength of Model 4 is about 1/2 that of Model 3.

(5) Model 5 (See Figure 7)

The improvement ratio is 20 percent and the bulkhead consists of three cells. The

cell is 120 cm in diameter, 91.5 cm in height and 0.27 mm in thickness. It has a 6 cm wide

footing but no stiffener.

Procedures for Making Models

Models are constructed right in the loading box of the shaking table. The relationship between cohesiveness and water content of untreated soil, and the relationship among

the cohesion, water content, and cement content of treated soil are determined by preliminary laboratory tests. Following these findings, the prescribed amounts of clay, water,

and cement are mixed and model ground of expected strength can be made. About 32 m^ of

soil is necessary to make the model ground, which must be made accurately and in a short

time. Therefore, both construction procedure and soil quality is controlled carefully.

Clay is mixed with water and cement by mixer (capacity: about 0.7 m^). Clayey grounds are

made with a clay of prescribed water content in each batch. Improved ground is made with a

clay, water and cement. Physical properties of clay are shown in Table 3.

In the case of pile models, plastic pipes are set up and fresh treated soil is poured

into them. Then, untreated soil is placed around the piles, and the plastic pipes are

pulled out. The grounds are not consolidated.

171
Five model caissons are placed in the direction of the face line of the bulkhead.

Openings of 1 to 2 cm are left between the caissons so they can move independently. Clear

sand is put in as backfill, and water is poured into the loading box up to the low water

level.

Shaking Conditions

The shaking table at the Port and Harbour Research Institute, Ministry of Transport,

which is shown in Figure 8 is used. This is a free vibration type shaking table as illustrated in Figure 9 [3]. It generates a free vibration in one horizontal direction, which

differs from the usual ground movements during earthquakes. Examples of the table and

model acceleration records are shown in Figure 10. Every model is shaken 5 to 6 times

with 50 to 400 gals or more of maximum acceleration. Predominant frequencies are 1 to 6 Hz,

within the range of low frequencies of the design earthquake conditions; consequently the

test vibrations are considered destructive. However, the number of predominant waves is

only 2 to 5 which is rather few compared to the design earthquake condition.

Measurements

As shown in the model cross sections, accelerations, earth pressures, and ground

strains are measured. Accelerations are measured primarily at the bottom of loading box

(table acceleration), at the surface of the clayey ground, improved ground, mound, backfill

sand, and at the top of caisson. Earth pressures acting from the backfill to the wall, and

the pressures beneath the caisson are measured. In order to observe the occurrence of

sliding failure, flexible plastic bars with strain gauges (strain meter) are buried in the

ground and their bending strains are measured.

EXPERIMENTAL RESULTS

Accelerations of Ground and Wall

As acceleration examples, the first and the second peaks of several points are shown

in Figure 11 and 12. In the Model 1 results shown in Figure 11, the improved ground is

rather rigid and the model accelerations are amplified to less than 600 gals of the table

acceleration. The caisson accelerations are not as large because the gravel mount has a

shock absorbing effect against the severe disturbances. Rocking vibrations occur to the

172
caisson. Figure 12 shows that the improved pile ground reacts rather flexibly to large

accelerations. Figures 13 and 14 show the ratios of first peak acceleration of Models 2

and 5 to that of the shaking table. The influence of average strength and the type of

improved ground is found from the acceleration response.

Displacements of Wall

The horizontal displacement, settlement, and tilt of the caisson and cell walls were

measured after each increment of acceleration increase. As little settlement or tilt

occurred in the tests, the mounds and improved grounds are considered to have sufficent

bearing capacity. However, the horizontal wall displacement does increase with the

increase of table acceleration as shown in Figures 15 and 16. In Figure 15 the differences

due to the variations in the models are not clear from the relationship between wall displacement and table acceleration; but as shown in Figure 16, the walls of Model 1 and

Model 2 start sliding at 200 to 300 gals of acceleration on the surface of the improved

ground, while the walls of Models 3, 4, and 5 are displaced by much smaller accelerations.

It was observed after the tests that the pile tops of Models 3, 4, and 5 were displaced

several centimeters horizontally. Consequently, it is felt that the wall displacements

are caused by the pile top features. The strain meters in the ground do not indicate any

signs of sliding failure.

Pressures Beneath Caisson

Dynamic pressures beneath the caissons are measured by earthpressure cells buried in

the mound. Examples are shown in Figure 17, where the static pressure is assumed to be the

sum of the weight of caisson and earth-pressure due to the backfill sand. The increase in

dynamic pressure indicates the occurrence of rocking vibrations on the caisson.

Damage of Backfill Sand

Settlement and cracks on the surface of the backfill sand were observed after each

shaking. They increase along with increases in the horizontal wall displacements and their

aspects are similar to the seismic damage of gravity type bulkheads.

173
DISCUSSION

The purposes of these tests are to provide an understanding of the characteristics of

seismic behavior and to collect the data to develop a numerical simulation method. The

complete investigation cannot be reported on in this report because the numerical simulation method is still under development and the tests with the five models have just been

completed. Therefore, the experimental results are compared to the results of the stability analysis using present procedure. On the other hand, the seismic stability of the

prototype is estimated in consideration of the similitude.

Stability Analyses by Present Design Procedure

According to current design procedure for harbor structures [4], the stability of

model walls against sliding is estimated as shown in Figure 18. Safety factors are calculated using a seismic coefficient, the ratio of maximum acceleration at the improved ground

surface to the acceleration of gravity. The arrows indicate the informal cases, in which

the square root is neglected since it becomes imaginary in calculating dynamic earthpressure. The face line swelling in Models 1 and 2 is considered to be caused by the

sliding of the walls. Since the walls start to slide at the safety factor of about 1.0,

estimates from the present design procedure do not contradict the experimental results.

In Models 3 to 5, it is considered that the face line swelling is caused by the displacement of the pile tops. The swelling starts about 1.8 or less of the safety factor.

Safety factors of circular failure plotted against horizontal wall displacements

are shown in Figure 19. The seismic coefficient is the ratio of table acceleration to

acceleration of gravity which distributes uniformly through the model. Since the face

line swellings in the block models are caused by the walls sliding on the mound, it is

better to exclude them from this discussion. In Models 3 to 5 the walls start to slide

at a safety factor of 2.5 or less. From observations of the improved ground, these

swellings are not caused by shear failure of the treated soil but by the displacement of

treated soil piles. Since rigidity and rupture strain are somewhat different between

treated and untreated soils, further investigation is required despite the large circular

failure safety factor.

174
Stability Prototype

The prototype is designed to a seismic coefficient of 0.2 by present design procedure.

From design seismic conditions at the construction site, the maximum acceleration at the

base and at the ground surface are 270 to 290 gals, and 200 to 290 gals, respectively.

Predominant incident wave frequency is 0.1 to 4.0 Hz. The number of predominant waves is

from 20 to 30.

Although there are some uncertainties, the particular tests are chosen as the

equivalent disturbances to design seismic conditions as shown in Table 5. The number of

predominant waves is somewhat less in the tests than in the design seismic condition, but

it is compensated for by the somewhat larger acceleration. Also, the predominant frequencies are within the range of low frequencies, and test vibrations are in the destructive

range

.

When the similitude between model and prototype is regarded as satisfied, the

stabilities of prototypes are estimated in Table 5. According to Table 5, sliding and

wall settlement takes place to some extent but ground and bulkhead failure does not occur.

Consequently, it is estimated that the prototype does not lose its function under the

design seismic condition.

CONCLUDING REMARKS

In order to rationalize earthquake resistant design procedure for bulkheads on ground

improved by the D.M. method, much research such as the model tests, earthquake observations, and numerical simulations have been actively pursued. This paper reports only the

outline of the shaking table tests on relatively large models, hence the integration of the

related research and further investigations are required.

REFERENCES

[1] Terashi, M. , "Chemical Soil Stabilization Methods Applicable to Port and Harbor

Construction," Proc . of 1977 Annual Research Presentations of Port and Harbor

Research Institute, pp. 63-100, December 1977.

[2] Nakaraura, R. , "Deep Mixing Method by Cement Slurry - (I) and (II)," Umetate to Shunsetsu, No. 78, pp. 33-55, 1977 and No. 79, pp. 23-38, 1978.

[3] Chiba, T. and Noda, S., "On the Free Vibration Type Shaking Table," Proc of 34th

Annual Convention, Japan Society of Civil Engineers, Vol. 1, pp. 426-427, 1979.

[4] The Japan Port and Harbor Association, "Engineering Requirements on Port and Harbor

Facilities and Its Application," March 1979.

175
Table 1 Similitude

Parameters Symbols Ratios

Length Lr L

r

= 1^,/Lp = 1/20

Unit Weight Y r Y r = YnAp = 1

Acceleration ar a r = Vap

= L r /T/ = 1

Time Tr T

r = Tm

/Tp = Lrl/2 = 1/4.47

Weight, Force wr Wr = w

m

/ wp

= Y rL

r

3 = 1/20 3

Cohesion c r C

r = Cm

/C

p

= W

r /Lr

2 = Yr L

r = 1/20

Angle of Internal

Friction

*r * = *m/*p

1

m : Model, p : Prototype, r : Ratio of Model to Prototype

Table 2 Model Details

Model 1 Model 2 Model 3 Model 4 Model 5 Wall Type Caisson Caisson Caisson Caisson Cell

Ground Type Block Bio. k Pile Pile Pile Improvement Improvement Ratio (%) 100 100 10 20 20

Wall

Height (cm) 85 102 85 85 91.5 Width (cm) 55 55 55 55 120 Backfill Sand (cm) 115 115 115 115 1 1 r

> Layer Mound (cm) 40 25 40 40 16 Thickness Ground (cm) 100 inn 100 100 1 no Sand (cm) 20 10 11! 10 in Water Depth (cm) -80.0 -87.5 -87.5 -87.5 Clay Used Kawasaki-A Kawasaki-A Kawasaki-B Kawasaki-B Kawasaki-B

Clayey Water Content W (%) 90 90 100 105 105 Ground Cohesion CQ (kg/cm 2

) 0.0104 0.0113 0.0160 0.0160 0.0140 Water Content W.( (%) 200 .Mill 200 200 200 Treated Soil Cement Content A„ (%) 10 8 15 11 1 1 Cohesion (kg/cm 2

) 0.635 0.370 1.640 1.018 0.843 Composite

Ground Cohesion C2 (kg/cm 2

) (0.635) (0. 370) 0.178 0.216 0.180

176
Table 3 Physical Properties of Clay

Specific Gravity of Solid G s

Kawasaki-A

2.69

Kawasaki-B

2.64

Consistency

Liquid Limit WL (%) 68.8 84.2

Plastic Limit Wp

(%) 33.3 34.5

Plasticity Index Ip 35.5 49.7

Grain Size

Distribution

Gravel (%) 0 0

Sand (%) 6.8 4.4

Silt (%) 47.1 38.6

Clay (%) 46.1 57.0

Table 4 Specifications of Shaking Table

Max. Horizontal Acceleration 1 .2 G (Full Loading)

Max. Horizontal Displacement ± 90 mm

Max. Horizontal Velocity 102 cm/sec

Frequency 0.8 ~ 4.4 Hz (Full Loading)

Loading Capacity 130 tf

Loading Box Length 8 m x Width 4 m x Hight 2 m

Table 5 Estimated Behaviour of Prototype

Design Seismic Condition Model 1 Model 2 Model 3 Model 4 Model 5

Max. Acceleration of Ground (gal) 200 - 290 612 429 283 250 196

Max. Acceleration of Base (gal) 270 - 290 562 372 462 457 474

Predominant Frequency (Hz) 0.1- 4.0 0.52 0.74 0.48 0.41 0. 34

Number of Predominant Waves 20- 30 4 3 2 2 2

Max. Acceleration of Wall (gal) 571 348 297 212 214

Horizontal Displacement of Wall (cm) 58 38 76 44 52

Settlement of Wall (cm) 22 36 6 2

177
Fig.l Flow Chart of Development of Earthquake Resistant Design Procedure for

Bulkheads on Improved Ground (Ministry of Transport)

Shaking table test

'Sine wave test to \

,middle scale model/

Free vibration

test to large

scale model

Earthquake observation

Observations at Yokohama and Hiroshima ports

Numerical simulation

analysis

/ Development of non-linear analysis

\ program

Dynamic test of soil

Dynamic triaxial

compression test, Resonant cloumn

test ,

Dynamic characteristics

of model

Failure characteristics of model

Stress condition and ultimate stability of model

Seismic behaviour of prototype

Stress condition of prototype

Improvement oi program

Stress condition and ultimate stability of numerical model

Establishment of rational earthquake

resistant design

procedure

Dynamic properties of clay and treated soil

178
Width : 4m , 1 FT")

Fig. 2 Cross Section of Model 1

$ Accelerometer

Width : 4m . 1 fTl

Fig. 3 Cross Section of Model 2

$ Accelerometer

Earthpressure Cell

Width : 4m . 1 ITI

Fig. 4 Cross Section of Model 3

179
0 Accelerometer

Earthpressure Cell ^ Displacement Meter

Width : 4m Fig. 5 Cross Section of Model 41 m

0 0 0 c o 0 0 0 0 0 0 0 0 0 0 0

o o o o o o o o o o 0 o o o o o

o o o 0 o o o o o o o o o o o o

o o o c 0 o o o 0 o 0 o o o o o

o o o o o 0 o o o o 0 o o o o o

o o o o o o o 0 0 0 0 0 o o 0 o

o o o o o o o o o 0 o 0 o o o o

o o o c o o o o o o o c o o o o

o o 0 c o o o o o o o o o o o o

o o o c o o o o o o o o o o o o

o o o c o o o 0 o o o o o o o o

o o o c o o o 0 o o 0 o o o o

o c o 0 o o o o o o o o o o4

Clayey Ground f J3J5 Improved Ground

1 m ,

Fig. 6 Plane of ground (Model 4)

9 Accelerometer

Earthpressure Cell ^ Displacement Meter _2

Clayey Ground

i Improved Ground (Pile Type)^

Width : 4m 1 m

Fig. 7 Cross Section of Model 5

180
Fig. 8 Free Vibration Type Shaking Table

A/WV

. rrSpring

Oil Pressure Piston

Rail O O Q O

Wall

Fig. 9 Mechanism of Shaking Table

qal Model 4 400. r Top of Caisson

Fig. 10 Acceleration Records

181
Model 5

t

I I I I I I I I 1 I I -1000 -800 -600 -400 -200 0 200 A 00 600 800 1000

Acceleration (gal

)

Fig. 12 1st and 2nd Peak Accelerations (Model 5)

182
Model 2 O O T °P ° f Caisson A A Surface of Mound Q——n Surface of Improved Ground

I I I I I I I I I I I I 0 200 400 600 800 1000

Shaking Table Acceleration (gal)

Fig. 13 Ratio of Acceleration to Shaking Table Acceleration

Fig. 14 Ratio of Acceleration to Shaking Table Acceleration

183
Shaking Table Acceleration (gal)

Fig. 15 Horizontal Displacement of Wall to Shaking Table Acceleration

Fig. 16 Horizontal Displacement of Wall to Acceleration at Surface of Improved Ground

184
A Test 3

(kg/cm 2

)

Fig. 17 Dynamic Pressure beneath Caisson

10.0r

^

O Model I A Model 2 —& Model 3 • Model 4

Model 5 §50h Seismic Coefficient : Q= -yr- —A

'

—O —

1 O

AQ

A ' a ^'t" Q—1— ®

1-0 "2-0 3-0

Safety Factor of Sliding

Fig. 18 Horizontal Displacement of Wall to Safety Factor of Sliding

185
15-Qr

O Model 1 O A Model 2

Model 3 • Model 4

;10 0h a

A Hode ' 5

Seismic Coefficient : P- SF9

t 50-

B <tf 4h O O 1 O—

i

1 0 2 0 3 0 40 5 0 6 0

Safety Factor of Circular Failure

Fig. 19 Horizontal Displacement of Caisson to Safety Factor of Circular Failure

186
FULL-SCALE EXPERIMENTAL STUDY ON THE ASEISMIC PERFORMANCE

OF MEDIUM-RISE REINFORCED CONCRETE WALL STRUCTURES

by

M. Hirosawa

T. Goto

H. Hiraishi

M. Yoshimura

ABSTRACT

This study reports on an investigation of the seismic safety of medium-rise apartment

houses of RC wall construction through a full scale testing program. In Japan more than

700,000 apartment houses, mainly within five stories, have been constructed. Their seismic

safety, through structural experiments and studies of earthquake damage, has been confirmed

primarily because of their strength in lateral resistance which compensates for an

insufficient deformation capacity.

Recently, demand for eight story construction of this type has increased due to

factors such as the increased cost of building sites, and the advantages of such structures

in regard to urban planning. Accordingly, in 1979 an outline of a new design Code for RC

wall structures was written.

In the new design Code the adopted principle is to keep ductile deformation capacity

to compensate for less strength compared to that of past RC wall structure within five

stories. To this purpose, the following several regulations to improve structural design

deformation capacity were provided: to have shear walls with uniform stiffness, to

restrict the number of members which could be anticipated to fail from brittleness, and to

set a minimum value for the shear reinforcing ratio.

Tests of a model structure, designed using the new Code, were conducted. From these

tests, it was found that the ultimate strength of the building was more than 50 percent of

its total weight, and that the deformation capacity was sufficient.

187
OBJECTIVE OF TESTING

Japanese apartment houses of RC wall structure are designed according to the 1952

A.I.J. Building Code. This Code prescribes that cracks may not occur in specimens

subjected to a lateral force corresponding to a design base shear, k, of 0.2. To obtain

such strength, a required area, a width of shear wall, and a required shear reinforcing

ratio to prevent the progress of cracks were specified. Large-scale seismic tests on RC

wall structures were carried out. The tests showed such structures to have a sufficient

lateral strength which is equivalent to the vertical weight, even though they were inclined

to fail due to brittleness. Apartment houses of this type have hardly suffered significant

damage from the several earthquakes they have experienced.

The new Building Code for medium-rise RC wall structures aims at a structure with

increased ductile deformation capacity to compensate for the lowered strength from the

lesser area of shear walls.

The following fundamental criteria involving assurance of seismic safety against

severe earthquakes are adopted by the new Building Code: (see Figure 1 and Figure 2)

(1) To control stress to any part of the structure caused by f requently-occuring

medium strength earthquakes within permissible stress levels;

(2) To prevent failure to the structure from the rare, severe earthquakes.

The provisions for achieving these purposes are:

(1) Lateral resistant strength should be more than 50 percent of the total weight

of the structure.

In evaluating ultimate strength, the effect of transverse members such as slabs and

transverse walls, which cause increased flexure, and the shear strength of the longitudinal

members, may be included.

(2) The structure should be strong enough to not lose its strength under the

compulsory displacement of more than twice the yielding displacement.

For this purpose, the following is prescribed:

(a) The cross-sectional area of shear walls in each frame should be equalized, and

the structure should be designed to be as ductile as possible.

(b) Restrict the number of members which are prone to fail in nonductile manner,

i.e., deep beams, members with non-structural walls, and members with openings.

188
(c) To design members using the criteria that flexure failure should occur prior to

shear failure, and shear reinforcement ratio exceeds the specified value.

The test specimen reported here was designed to satisfy the above new regulations.

The following questions summarize what should be confirmed through testing:

(1) Regarding ultimate strength,

(a) Does ultimate strength exceed 50 percent of its own weight?

(b) To what extent are transverse members more effective than longitudinal

ones?

(c) Can the resulting ultimate strength be explained theoretically?

(2) Regarding the hysteretic loop,

(a) Is strength reduced when the displacement range goes beyond maximum

strength?

(b) Is there any which can be expected to fail from excessive brittleness?

If there is, does it not have a bad influence upon the total behavior of the

structure?

(c) Do members with non-structural walls, and with openings, have stable

behavior?

(d) How is the aseismic performance of the test structure estimated, on the

basis of ultimate strength and deformation capacity, through testing?

TEST SPECIMEN

The Choice of a Model Test Specimen

The main objective of this testing was to investigate the properties of the

provisions for seismic safety given in the newly established design Code outline. The

most essential factor to seismic safety in a wall structure is thought to be the shear

wall ratio, the total shear wall cross-sectional area divided by the floor area. Therefore, a structure having a shear wall ratio just beyond the regulated minimum values was

chosen for the test structure.

As for loading, it was limited in one horizontal direction because of the restricted

capacity of the full-scale testing facilities. The question of whether a longitudinal or

transverse direction should be chosen was examined. In the transverse direction, the

189
structure was composed mainly of rather slender shear walls, with an average shear stress

due to lateral force, 50 percent of its weight, of 15 kg/cm^, which was less than 10 percent of the concrete compressive strength. This implied that in the transverse direction,

the structure would fail in flexure prior to shear, which suggests that sufficient seismic

safety will be obtained. In the longitudinal direction, there was a mixture of several

shear wall types with different stiffness, so that the structure would not fail in flexure.

Accordingly, the longitudinal direction of the structure was chosen as a testing

objective

.

Further, the restricted funds and the necessity of maintaining the specimen in

non-optimum advantageous condition forced in other design changes as follows:

(1) Test specimen reduced to first five stories from the original eight story

structure

.

(2) Test specimen reduced to two apartment houses, a smaller part of the original

model structure.

(3) The dimensions of the transverse direction were reduced to 80 percent.

(4) Main bars in the beams of the sixth to eighth stories of the original structure

were moved to the beams of the fourth and fifth floor of the specimen to make the lateral

capacity of the specimen equal to that of the eight story structure.

(5) Slab reinforcement in the longitudinal direction, and shear reinforcement in the

transverse wall, remain the same as the original eight story structure.

(6) The effects of deep beams, small openings [7], and a spandrel wall [8] which the

original structure did not include, were added to the investigation.

(7) The effect of a shortened main bar anchoring length in the beams going into the

shear walls, which differs from the plans for the original structure, was also added to

investigation

.

Outline of the Test Specimen

The plan and sections of the test specimen are shown in Figure 3. Figure 4 shows one

example of reinforcement. Steel used in the specimen were all deformed bars, SD30, having

a nominal yielding stress of 3.0 t/cm^. The concrete was a normal one with design strength

of 240 kg/cm^, having a tested compressive strength of 240-280 kg/cm^.

190
LOADING AND MEASURING PROGRAM

The test specimen was fabricated on the floor of the full-scale testing facilities in

the Building Research Institute, Ministry of Construction. Testing lasted about one month

in November, 1979.

Prior to lateral loading, a vertical load of 840 ton, the equivalent of the weight of

the top three stories of the original eight story structure was applied with 16 oil jacks,

and was kept constant during the lateral loading. Lateral loading was conducted by 10, 6,

and 2 oil jacks at the roof, fifth and fourth floor levels, respectively. The overturning

moment at the first floor caused by this lateral load was equivalent to that of the model

eight story structure subjected to the uniform distribution of lateral load. Thirteen

times loading reversals were conducted under the control of rotational angle, R, mainly of

1/400, 1/200 and 1/100 ratios.

The following test data were measured automatically: lateral and vertical displacements of 140, steel strain of 328, strain on the concrete surface of 135, and rotational

angle at nodal points of 18

.

Additionally, cracks were observed in all members including transverse walls and

slabs. Significantly developed cracks were measured at each loading step. The loading and

measurement conditions, and the loading program are shown in Figure 5 and 6, respectively.

EXPERIMENTAL RESULTS

Outline of Testing Results

Table 1 shows the maximum value of loads and associated displacements at each loading

cycle. The relationship between load and absolute lateral displacement at roof floor is

shown in Figure 7

.

As is shown in Table 1, the maximum load was about four times the design load, and the

associated displacement at the top floor was 1/100 of the structure's total height which

was 27 times the elastic displacement of the design load. In subsequent loading, when displacement reached 1/50 of the total height of structure, the maximum load was 490 tons less

than the previous maximum load

.

191
Behavior of Specimen

Specimen behavior prior to obtaining maximum strength is summarized as follows:

(1) First cycle with half the design load showed hardly any cracks.

(2) Second cycle with the design load showed a few flexural cracks in the beams,

diagonal shear cracks at the end of openings, and flexural cracks in the slabs, but no

stiffness deterioration.

(3) At the rotational angle, R, of 1/800, besides progressive flexural and flexuralshear cracks in the beams and walls, horizontal cracks occurred to the transverse wall and

shear cracks to short-span beams and small-depth walls in the Y2 frame. The maximum load

was twice the design load and the associated displacement was 3.5 times the design

displacement

.

(4) At the rotational angle, R, of 1/400, (fourth cycle to sixth cycle), shear cracks

in the beams and walls and horizontal cracks to the slabs and transverse wall were progressed noticeably, while diagonal cracks occurred in the beam-column connections. Yielding in the beam and wall barely occurred. The maximum load and the associated displacement

were, respectively, 2.5 and 7 times the design values. Deterioration of maximum load due

to loading reversals was not observed.

(5) At R = 1/200 (seventh to ninth cycle), a yielding of the beams and walls began,

while shear failure of short-span beams and small-depth walls in Y2 frame became obvious.

The maximum load and the associated displacement were 3.5 times and 1.5 times the design

values. No deterioration of the maximum strength in loading reversals was observed.

(6) At R = 1/100 (tenth cycle), load reached its maximum. Yielding hinges were seen

in almost all the member ends, the short-span beams failed in shear. This implies pronounced deterioration of the maximum strength. Figure 8 shows the development of cracks

in the Y2 frame at each loading level.

Maximum Strength

The maximum strength was 1,608 tons in tenth cycle (R = 1/100). This corresponds to

86 percent of the total weight of the original eight story structure. The maximum strength

of the negative direction loading was reduced to 1,529 tons which is equal to 95 percent of

the 1,608 ton load. This indicates that the different width of the transverse walls under

the tensile stress due to lateral load does had some influence.

192
Failure Mechanism and Strength Reduction

Judging from strain gages attached to the longitudinal reinforcement and from the

developing crack widths on beam and wall ends, flexural yielding at the ends and the bottom

of first floor walls occurred at the rotational angle, R, of 1/400. At R = 1/200, ninetyone joints, that is, more than 65 percent of all joints, yielded by flexure. Until this

stage, only two walls and two beams of Y2 frame had failed in shear. At the maximum

strength (R = 1/100), joint yielding increased. Shear failure, mainly that of short beams

with 1.0 m span-length, also developed.

These test results indicate that it is appropriate to consider the failure mode of

this specimen as a flexural failure prior to a shear failure. Existing conditions when

hinging occurred were obtained by steel strain measurements and are given in Figure 9.

When the specimen was subjected to loading reversals three times under the control of

the same displacement after the maximum strength, 1608 tons, the maximum capacity was

reduced to 1,460 tons; and in the negative loading, the maximum strength was -1529 tons at

the same, tenth, cycle as positive loading. Load capacities for the twelfth and thirteenth

cycles were -1,368 tons and -1,347 tons, respectively. As shown above, strength reduction

was not pronounced and the displacement limitation for ductile behavior could be thought

to be 1/50.

The mean shearing stress, calculated by dividing the maximum strength, 1,608 tons, by

the totaled cross-sectional wall area is 23 kg/cm^, which corresponds to 1/10 of the compressive strength of the concrete test cylinders. This value of 23 kg/cm^ remains in the

region were satisfactorily repairable conditions are expected to occur when appropriate

reinforcement is in place.

Other Test Results

(1) Effects of transverse members

(a) Judging from the reinforcement strain and from the developing cracks in the

slab, it seems that the slab reinforcements are effective in increasing the flexural

strength of the beams at the stage of maximum strength.

(b) Slab concrete has some effect on the shear strength of beams because some

beams which failed in shear were associated with shear fracture of the slab.

193
(c) As for the effect of the transverse wall on the flexural strength of the

wall, it seems from the strain gage data that all the reinforcement in the transverse wall is effective, where the transverse wall is attached to the edge of the

longitudinal wall. It is also recognized, judging from the horizontal tensile

cracks in the transverse wall, that attaching the transverse wall to the center

of the wall increases the strength of the longitudinal wall.

(d) As for the effect of the transverse wall on wall shear strength, judging

from vertical tensile cracks occurring at the transverse wall from high compression

under large displacement loading, it seems that a part of the transverse wall transferred shear and increased wall shear strength.

(2) Small wall openings

Some of the walls have small openings which may be ignored in the design Code if the

reinforcement is properly arranged. At four corners of these openings, diagonal shear

cracks occurred in an earlier stage of loading when the mean shearing stress was about 2.5

kg/cm^. These cracks developed as the load was increased, but the widths did not expand

even in subsequent loading. In addition, crushing did not occur. Of twelve reinforced

openings which had strain gages attached, only two yielded. These results indicate that

there is no problem for reinforcement at small openings.

(3) Behaviors of beams with thin spandrel wall

Outer beams for the first and fifth floors of the Y2 frame had a thin spandrel wall

separated from an adjacent wall by a slit, which was deemed to be structurally negligible

according to the design Code outline.

Testing shows, the spandrel wall had flexural cracks near the slit at a lower load

about equal to 1/3 of the maximum load, which progressed to mid-span in subsequent loading,

but almost no shear cracks. Crushing at the top of the slit began near the maximum load,

and concrete in the circumference exfoliated at the maximum load. Beams with spandrel wall

and the walls that the spandrel wall joined did not show brittleness failure, which

indicates that the spandrel wall would not adversely influence surrounding members.

(4) Short-span beam behavior

Among ten beams of 1.0 m span-length and ten beams of 1.6 m span-length in the Y2

frame, eight beams on the second, third, and roof floors obviously failed in shear

194
after flexure yielding. Seven of eight beams having a wider cross-sectional area on the

fourth and fifth floor had slight damage. As a whole, this specimen was considered as a

flexure-failure frame except for a few beams and walls, while strength capacity did not

markedly deterioriate after exposure to maximum load.

(5) Failure of short-span walls in Y2 frame

Fourth and fifth floor Y2 frame short-span walls failed in shear. While these

fourth and fifth floor walls developed shear cracks at low levels of load, those of first,

second, and third floor had few shear cracks, and they did not notably progress. The

fourth and fifth floor walls had extended shear cracks at the rotational angle, R, of

1/200, and the concrete began to exfoliate at 1/100 R. Finally, at R of 1/50, exfoliation

of concrete was so extensive that the other side could be seen through the shear reinforcement .

The reason for failure is thought to be that these short-span walls had a "column"

deformation condition, and the adjacent walls with high stiffness had a "canti-lever"

deformation, with the result that shear forces were concentrated on the short-span walls

in the upper stories.

ANALYTICAL STUDIES ON TESTING RESULTS

Elastic Analysis

The general-purpose program for elastic analysis, TABS, by the University of

California, was used to investigate the behavior of structural members in the elastic

range subjected to the maximum test load, 1,608 tons.

The following assumptions were introduced:

(1) Structural members were treated as line members. Columns with rigid end zones,

including flexural, shear and axial deformation, and beams with rigid end zones, including

flexural and shear deformation, were considered.

(2) No torsion was thought to occur under the assumption of a rigid floor.

(3) With interaction between transverse wall and column, and slab to beam having the

flange width up to the mid-span in the transverse direction, all cross-sectional areas of

the transverse members could be estimated in calculating the stiffness of the structure in

the longitudinal direction.

195
(4) Despite assumption (3) above, shear area was considered as a cross-sectional

area of the longitudinal members.

(5) The column rigid zone length was estimated from the nodal point to the face of

the connecting beam, while the beam (rigid zone length) was from the nodal point to the

face of the connecting column.

(6) Story height was measured from floor to floor.

Detailed results for the elastic analysis were abbreviated in order to be shown.

Except that the calculated specimen stiffness was considerably higher than the measured

one, in general, the analytical results for stress distribution of each member sufficiently explain the conditions of cracking and yielding of each member in the test

observations.

Bearing Capacity

For the six cases shown in Table 2, the maximum specimen strength was analyzed using

the principle of virtual work. The test structures failed in flexure, i.e., the beams of

all floors yielded in flexure, and the walls yielded in flexure at the base of the first

floor. The calculated results of each case are shown in Figure 10 as well as the shear

force at the first floor.

The following can be observed from the figure.

(1) The calculated results are in good agreement with the test results in both the

positive and negative loading direction. However, it should be noted carefully that the

test and calculated results include the additional capacity due to the axial forces in the

beams

.

(2) The calculated results, which take into consideration the effect of all the

members being perpendicular to the beams and walls in the loading direction, agree well

with the test results. The ratios of the calculated strengths, considering no effect of

the perpendicular members, and then considering the effect of only half of the perpendicular members, to the above-mentioned calculated strength were 57 percent and 83 percent,

respectively. Accordingly, the effect of the perpendicular members to the bearing capacity

of the specimen was found to be remarkable.

(3) The result calculated using the nominal strength as the material strength

(3.0 t/cm^) for the yield strength of steel and 210 kg/cm^ for concrete compressive

196
strength) was 84 percent of the calculated strength using the test results as the material

strength (around 3.6 t/cm^ and 240 kg/cm^). This ratio almost corresponds to the ratio of

the nominal yield strength of steel to the actual yield strength.

(4) The additional strength due to the axial force in beam was determined to be

240 tons. This value corresponds to about 24 percent of the measured maximum strength.

(5) The calculated result, based on design Code regulations was 87 percent of the

test result.

Statistical Non-Linear Analysis

The method used for static non-linear analysis is as follows:

(1) Estimate the deteriorated stiffness of each member due to the occurrence of

cracking and yielding under an increasing horizontal load, based on the elasto-plastic

behavior of each member.

(2) Assemble each time the determined stiffness matrix of each member to the total

stiffness matrix of the structure.

(3) Calculate the increment of displacement under the increment of the external

force.

Fundamentally, since no change in stiffness is assumed during an increment of force,

this method may be said to be a repetition of the ordinary elastic direct stiffness method

Horizontal displacement at each story and rotational angle at each nodal point are considered as unknown factors, and axial displacements of walls and beams are neglected. The

Degrading Tri-linear model, which agrees with the hysteretic loop of reinforced concrete

members failing in flexure, is used as the hysteresis characteristic of the members. Further, shear deformation is included as a flexural one under the assumption that shear

rigidity deterioriates proportionally to flexure rigidity. Hence, shear collapse of the

members is not considered in the analysis. The calculated load-deflection relationship

at the roof and the second floor obtained by this method is shown in Figure 11, accompanie

by the experimental results. As shown in the figure, the calculated results agree fairly

well with the test results.

197
Non-Linear Dynamic Response Analysis of the Model Building

Non-Linear dynamic analysis was carried out on the eight-story building, which is

considered to be the original model for the test specimen using the following assumptions:

(1) The yield capacity of each story in the model building is calculated with the

conditions that the external force distribution is inverse-triangle, and that the overturning moment at the base of the first story due to the external force is equal to that

of the experimental results

.

(2) The total weight of the model building is assumed to be the same as the test

specimen, 1,879 tons. Weight of each story is uniform.

(3) The cross-sectional area of each member of the model building is assumed to be

the same as the member on the first story of the test specimen. Member stiffness is estimated taking into consideration the full extent of the transverse members.

(4) The stiffness reduction coefficient after cracking is assumed to be 5 percent of

the calculated elastic story stiffness, refering to the test results. The story cracking

strength is assumed to be 50 percent of the calculated story yielding strength.

(5) Damping is estimated at about 7 percent for the fundamental period within the

elastic range.

The results of the response analysis for the N-S component of the El-Centro Earthquake

(max. acceleration: 300, 500 and 700 gal) are shown in Figure 12. The figure shows that

response displacements in all cases are within a permissible range, and that the

eight-story model building exhibits excellent seismic performance.

CONCLUSIONS

A full-scale test on the first five stories of the original eight story apartment

house of reinforced concrete wall construction was carried out. From the experimental and

analytical results, the following was concluded.

(1) Bearing Capacity

The obtained maximum horizontal capacity of the specimen was 1,608 tons at the base

floor. This value corresponds to 86 percent of the total weight of the model building.

However, neglecting the additional strength due to axial stress in the beams, this value

is reduced to 1,365 tons or 73 percent of the total weight.

198
The calculated maximum strength, using the assumption that the failure mechanism is

a beam-yield type and that wall and floor slabs perpendicular to the walls and beams in the

loading direction are effective over the entire length, coincided closely with the test

results.

(2) Deformation Capacity

Strength reduction was not significant even under a large displacement which

corresponds to a rotational angle of 1/50: excellent deformation capacity was recognized.

The elastic and plastic characteristics of the structure could be analyzed using the

postulated assumptions.

(3) Synthetic Estimation of Seismic Performance of the Model Building

The foregoing results on bearing capacity and deformation capacity satisfy the

regulations of the design Code. Consequently, the seismic performance of the model building test specimen is considered excellent. This is also shown by the results of dynamic

response analysis using several strong earthquake motions.

(4) Notes on the Structural Design

(a) The extensive width of the perpendicular members was effective in providing

flexural strength over the entire length of this specimen.

(b) The reinforcing used around the small openings in the wall was recognized

to be sufficient.

(c) Thin spandrel walls, which were not considered in the structural

calculation, did not unfavorably effect the beams and walls connected to them.

(d) Brittleness failure was significantly recognized in several short-span

beams. However, since the number of members failing due to brittleness was small,

the influence of a partial brittleness failure on the whole structure was found to

be negligible.

199
Table 1 A List of Max. Total Load and Measured Horizontal

Displacement at each Story in each Loading Cycle

(Unit: Load-ton, Di spl acement-mm)

Loading

Cycl e

Total Load Horizontal Displacement (Yq Plane} Note

Scheduled Story

Deflection Angle

6F 5F 4F 3F 2F IF

1 + 198 2.61 2.24 1 .81 1.12 0.39 0.05 1/2 times Design

198 3.07 2.64 1 .41 0.63 0.41 0.01 Load (k=0.2)

2 + 396 4.79 3.96 3.49 2.45 1.13 0.19 1 time Design

396 5.78 5.02 4.23 2.10 1 .06 0.05 Load

3 + 779.4 16.98 14.09 11 .47 7.69 3.59 0.49 (1/800)

759.0 16.85 14.69 11 .14 6.01 2.49 0,09

4 + 1112.4 35.16 29.90 25.30 16.85 8.19 1.17 No.l Cycle at 1054.0 33.93 28.94 22.55 11 .95 4.32 0.29 1/400

5 + 1048.4 35.60 30.64 26.30 17.61 8.64 1 .27 No. 2 Cy::e at 994.6 33.87 29.50 22.61 11 .88 4.19 0.34 1/400

6 + 1035.0 35.23 31 .07 26.50 17.96 8.77 1 .42 No. 3 Cycle at 960.2 33.83 29.17 22.28 11 .46 3.67 0.43 1/400

7 + 1444.0 71 .23 61 .54 53.01 37.07 18.39 3.41 No. 1 Cycle at 1322.0 66.00 56.94 44.42 22.58 8.06 1 .26 1/400

8 + 1209.6 71 .73 63. 17 52.89 36.07 18.60 3.73 No. 2 Cycle at

1291 .4 66.03 56.94 44.35 22.65 8.12 1 .27 1/400

9 + 1203.0 71 .53 63.00 53.41 36.49 18.81 3.88 No. 3 Cycle at

1251 .0 66.33 57.04 43.88 22.37 7.96 1 .30 1/400

10 + 1608.0 132.50 116.75 97.29 66.40 35.86 10.54 No.l Cycle at 1528.6 133.33 111 .89 82.73 39.35 16.43 5.66 1/100

11 + 1488.2 139.93 123.74 101 .99 71 .02 38.80 12.14 No. 2 Cycle at 1430.6 133.23 111 .82 84.28 44.60 16.55 6.28 1/100

12 + 1460.2 144.36 127.31 98.10 68.15 37.53 12.49 No. 3 Cycle at 1368.4 132.80 111 .42 84.41 44.74 16.53 8.26 1/100

13 + 1548.2 211 .29 186.38 155.81 107.72 56.33 19.47 Last Cycle at 1347.0 179.75 148.95 116.69 64.17 28.48 17.36 1/50

200
Table 2 A List of Structural Conditions Assumed in the Cases

of Maximum Capacity Calculation

Case Loading

Direction* 1

Considered

Extention of

Perpendicul ar

Wall

Material

Strength* 2

Axial Force in Beams* 3

1 Positive Zero Test Result Negl ected

2 Positive All Test Result Negl ected

3 Positive All Test Result Considered

4 Positive Half Nominal Negl ected

5 Positive All Test Result Neglected

6 Negative All Nominal Neglected

Note)

* 1 : Thick (thin) perpendicular wall is in tention side

in case of positive (negative) loading.

* 2 : Nominal yield stress for reinforcing steel is 3.0 t/cm 2 (SD30)

Design concrete compressive strength is 210 kg/cm 2 at 4 weeks.

*3 : Axial forced in the R 5 and 4th floor beams due to concentrated loading applied in the test.

201
0 5

~ 10

Deflection Angle ( R=^/H • x10

3 )

Fig. 1 Idealized Horizontal Load Vs. Deflection Relationship of Low- and Medium-Rise Reinforced Concrete Wall Structures

X a X

0 H 0 X

X 0 0

X

0 0 X

X 0 0

X a W

X 0 0

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0 0 E X

X 0 0H _v_

0 0 B X

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Section in case of

5 -Storied Building

—

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Change

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to Linear

Elements

j:

n r n n

>< l 1 X X ;X

X^\

'Thick

.Walls XX H s x^

\ / H

xx North-Side Section

of Test Specimen

Fig. 2 Change of Structural Systems in Low- and Medium-Rise Reinforced Concrete Wall Structures

202
Fig. 3 Plan and Structural Planes in the Loading Direction of the Specimen

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204
Vertical Load Corresponding to Dead and Live Load of Omitted

Upper 3 Stories

Horizontal

Displacement

MeasurementHorizontal Load

Corresponding to

Earthquake Force

Test Specimen

Lower 5-storied of 8-

stofied Model Building

Strain Measurement

of Steel and Concrete

Fixation of the Specimen

to Test Bed

FIG. 5 LOADING AND MEASURING SYSTEMS APPLIED TO THE TEST

FIG. 6 LOADING EXCURSION APPLIED TO THE TEST

205
Fig. 7 Base Shear Vs. Top-Story Displacement Relationship

206
a) After No. 2 Cycle

!

J. t. >> 1 -—

s

1

b) After No. 3 Cycle

II I!

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Fig. 8a. Observed Crack Pattern (No. 1)

207
c) After No. 4 Cycle

208
Fig. 8c. Observed Crack Pattern (No. 3) 209
i9 r-"-

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o Flexure Yield

at R = 1/400

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at R = 1/200

210
1500-

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' Fig. 10 Influence of Several Factors on Calculated Maximum Strength

211
212
SOIL DYNAMICS - STATE OF THE ART, 1980

by

W. F. Marcuson III

A. G. Franklin

P. F. Hadala

ABSTRACT

This paper describes and evaluates important modern developments in geotechnical

earthquake engineering. The current state of knowledge allows for the safe design and

construction of critical structures subjected to earthquake loadings, although we generally

do not know precisely the margin of safety that has been incorporated in the design. For

evaluation of the seismic stability of existing critical structures, our state of knowledge

is sometimes inadequate. In the analysis of existing structures we can define conditions

which are clearly safe, and conditions which are clearly unsafe. Between these two limits

there are many practical cases that fall into a grey area which will only be narrowed by

further research and new full-scale response data.

213
INTRODUCTION

The objective of this paper is to describe and evaluate important developments in soil

dynamics, particularly those of the last fifteen years. Primary emphasis will be given to

earthquake engineering, for three reasons: first, because earthquake engineering involves

the same fundamental issues and problems that affect the whole field of soil dynamics;

second, because of all areas of soil dynamics, earthquake engineering has seen the greatest

intensity of research and development effort in the past 15 years (for example, in 1975

approximately 30 million dollars was spent on earthquake-related research, on a world-wide

basis, as reported by Lee et al., (1978); and third, because the interests and experience

is primarily in the United States, this paper primarily reflects the state-of-the-art of

practice on the North American continent.

The categories into which soil dynamics problems are conventionally placed are often

based primarily on the nature of the loading function. Major categories of dynamic loading

include earthquakes, machine vibrations, pile driving, blasting vibrations, shock waves,

projectile impacts, and vehicular loadings. The differences among the various categories

of dynamic loading lie primarily in the frequency content or rate of loading that is characteristic of each one, and to a lesser extent on the stress levels or accelerations. We

consider soil response to machine vibrations to be essentially a solved problem, and other

state-of-the-art reports (Whitman and Richart, 1967; McNeill, 1969; Yoshimi, et al . , 1977;

Richart, 1978) cover much of this work. We will generally omit specific reference to all

types of dynamic loading other than those associated with earthquakes. We also will omit

reference to the prediction of ground motions, since this falls in the field of engineering

seismology rather than that of soil dynamics.

From a practical point of view, there are currently two approaches to obtaining

engineering solutions to geotechnical earthquake engineering problems: (1) an empirical

approach, and (2) a tuned analytical approach. Conceptually, the empirical approach consists of a systematic gathering of data on past performance and the organization of the

data in such a way that we have coherent patterns of behavior that can be used to predict

future performance. The method is essentially correlative but takes advantage of the

degree of understanding of cause and effect relationships that presently exists. By

"tuned analytical approach" we mean a method of predicting performance based on an

214
analytical model, where the results of past analyses with the model have been compared to

field case histories and correction factors have been developed to adjust the predicted

values to agree with those that were observed. The use of such an approach requires the

formulation of a workable theoretical model, and thus some understanding of the mechanisms

and processes involved. Imperfections of the model and/or systematic errors in the input

data are compensated for by correction factors. In the past 10-15 years extensive efforts

have been devoted to the development of analytical or tuned analytical approaches to

solving geotechnical earthquake engineering problems, but there has been a recent shift

toward a more equal balance between the empirical and analytical approaches.

MAJOR ISSUES IN SOIL DYNAMICS

We can identify as major issues in soil dynamics several areas that are of concern

because they constrain our ability to solve problems, or are areas of controversy, or are

particular foci of effort or attention. They are not all of equal importance, and indeed

we can see that the concentration of effort and progress in dealing with these issues has

been very uneven. These areas are treated in the following sections.

LIQUEFACTION

The fundamentals of the liquefaction mechanism are, at the present time, not

adequately understood, in the sense that we can not now formulate a model of material behavior that adequately describes liquefaction response and is amenable to numerical analysis

(Hardin, 1978). In the present state-of-the-art, we are analyzing liquefaction problems

using a mixture of theoretical concepts, empirical procedures, and hybrid procedures based

on tuned numerical analyses. The issue is a highly controversial one; far from having general agreement on the mechanisms of liquefaction, the profession sometimes cannot even

discuss the issue using a set of generally agreed upon definitions of terms (Marcuson

1978). We use the term "liquefaction" here in an inclusive sense to denote any of various

phenomena that involve high pore pressures, loss of shear strength, and excessive deformation in saturated cohesionless soils. Included are phenomena which have been variously

referred to in the literature as liquefaction , limited liquefaction, and cyclic mobility

(Castro 1975; Casagrande, 1976; Seed, 1979a).

215
At the present time we have available an abundance of laboratory data on the

liquefaction behavior of saturated clean sands and a limited number of well documented

case histories. Current effort is focused on the seismic performance of medium dense

sands, as the performance at the extremes of the density scale is easily predicted. However, our ability to predict the response of silty sands or gravelly sands is deficient.

This is partly due to a lack of understanding of the fundamentals of liquefaction and

partly because of limitations in our ability to perform adequate laboratory tests.

STRESS- STRAIN-TIME LAWS

These "laws" express mathematically whatever we know or hypothesize about the behavior

of a material. The stress-strain-time laws that we are presently able to make use of in

dynamic analyses are physically imperfect, but we do not find these imperfections to be

serious constraints on our ability to use analysis to obtain useful insights into engineering problems. The various mathematical models available in the present state-of-theart do not describe all aspects of soil behavior equally well (Hardin, 1978). Therefore,

obtaining a workable engineering solution to a problem depends on the selection of a mathematical model that best describes that aspect of behavior in which we are interested, and

selection of laboratory tests best suited to measuring the parameters of that model. For

instance, we must at the present time use different models in the solution of wave

propagation problems and the solution of dynamic strength problems.

Assuming that an appropriate model is used, a greater degree of uncertainty is

produced in the results by imperfect knowledge of the input paramaters than by defects in

the models themselves, so that the importance of model defects is relatively minor. The

input parameter values are obtained through field exploration and laboratory testing programs. To put the matter into perspective consider that most of the computer codes and

mathematical models that represent the current state-of-the-art were developed in the

1970' s (Lysmer, 1978); state-of-the-art laboratory tests were developed in the 1960's

(Woods, 1978), and state-of-the-art methods of obtaining undisturbed samples were developed in the 1940's (See State of the Art on Current Practice of Soil Sampling , 1979).

216
ANALYTICAL PROCEDURES

The development of analytical procedures is relatively far advanced, so that the

degree of sophistication that has been achieved is ahead of our present ability to provide

input data good enough to make full use of it. The late 1960's and the 1970' s have seen

the development of one-dimensional, two-dimensional, and three-dimensional methods of

analysis of dynamic stress wave propagation, using both linear and nonlinear stress-straintime laws as well as the equivalent linear approach, which involves some features of both.

These methods all involve the solution of the wave equation and their common objective is

to obtain time variation of shear stress within earth materials. The solution is accomplished by closed form, characteristic, finite element or finite difference methods. The

latter offers an advantage of greater flexibility in the use of sophisticated material

stress-strain laws (Lysmer, 1978).

Probably the most keenly felt limitation in current methods of analysis is that none

of them are capable of reliably predicting deformations, using either simple or complex

mathematical models. The shortcomings lie partly in the stress-strain-time laws, partly

in the problems of adequately determining moduli for tests on so called undisturbed samples, and partly in the analytical procedures themselves. Computation of dynamic displacements and deformations involves integration of accelerations, and progressive amplification

of relatively small errors in those values. On the other hand, stresses and accelerations

can be computed with fair reliability so long as relative stiffnesses and masses are known,

because these calculations rely primarily on a balance of forces, including inertial

forces

.

DETERMINATION OF SOIL PROPERTIES

Field Methods

In practice, field measurement of dynamic properties of soils means the measurement of

compression wave and shear wave velocities, from which dynamic modulus values (applicable

to a particular range of stress and strain levels) can be computed, using seismic exploration technique (Ballard & McLean, 1975). The past decade has seen considerable development

in geophysical instrumentation and field techniques, including seismic sources and detectors for use in boreholes, improvements in shear wave sources, improvements in

217
interpretation methods, and the use of "signal enhancement" which involves the use of a

series of impulses and the algebraic summing of the successive signals received at the

detector in order to improve the signal to noise ratio (Woods, 1978). These methods

involve the response of soils at very low strain levels (of the order of 10

-

^ percent) so

that the modulus values obtained approximate the initial tangent moduli. Some work has

been done in recent years on large-strain seismic velocity tests (Shannon & Wilson and

Agbabian & Associates, 1979), but this has not developed yet to the point where such tests

are in routine use. Figure 1 summarizes the shear strain amplitude capabilities of field

technique. The determination of variations in moduli with strain level now routinely

relies on supplementary tests performed in the laboratory. An additional limitation is

that there is at the present time no method for the measurement of material damping values

in the field, and this parameter is particularly critical in the analysis of soil-structure

interaction.

A sometimes worrisome problem is that seismic wave velocities measured by different

investigators show greater variations than one would expect in an apparently simple physical measurement. These inconsistencies are not easily explained, but most of the common

errors involve such things as unaccounted for detonator delays or borehole deviations, and

too much distance between source and receiver resulting in a travel path less direct than

the one assumed (i.e. refracted through adjacent high-velocity zones), so that the reported

velocity values are too high. Generally, refinements in wave velocity measurement

techniques result in lower velocities being measured.

Table 1 summarizes the current techniques for measuring in situ dynamic soil

properties

.

With regard to evaluation of liquefaction potential, field testing methods are

generally not applicable to the direct determination of soil parameters, other than density, related to liquefaction behavior. The Standard Penetration Test (SPT) is used in

empirical evaluation of liquefaction potential, and other field tests such as the cone

penetrometer test have the potential for being used in the same way. The difficulties

that we encounter in evaluating liquefaction potential through laboratory tests tempt us

to look to in situ testing methods for a remedy. However, a general limitation in this

approach is our inability to control or even to know the in situ state of stress, the

218
drainage condition, and the volume or mass of material involved. Consequently, most in

situ tests are usable for this purpose only as index tests and will rely on correlation

with laboratory response and/or field observations of earthquake effects.

Laboratory Methods

In the current state-of-the-art, production testing uses the stress-controlled cyclic

triaxial test as a liquefaction test and the strain controlled cyclic triaxial test or the

resonant column test to measure modulus and damping. It is generally recognized that the

stress-controlled cyclic triaxial test has serious shortcomings in that (a) it does not

correctly represent the state of stress that we believe or assume to exist in the field and

(b) it involves reversals of the principal stresses that we do not believe occur in the

field. Other possiblities have been and are being explored, such as large-scale shake table

tests, hollow cylinder torsional tests, and simple shear box tests. All of these alternatives have problems that have not yet been overcome, such as difficulties in preparing specimens, difficulties in applying shearing forces at the specimen boundaries, and

nonuniformity of strain distribution.

One of the most troublesome problems in laboratory testing of soils is that of sample

disturbance, particularly in sands. Relatively recent research has shown that structure is

much more important in the behavior of sands than had previously been thought, and that

sand samples compacted to a given density in the laboratory can have very different structures and very different cyclic strength response, depending on the method of sample preparation (Ladd, 1974; Mulilis, Chan and Seed, 1975; Marcuson and Townsend, 1976; Mori, Seed

and Chan, 1978). These circumstances make the use of undisturbed samples imperative, but

the field sampling technology with which we are presently working represents the state-ofthe-art of the 1940 's (Hvorslev, 1949; American Society of Civil Engineers; 1978; Marcuson

and Franklin, 1979; Horn, 1979).

On a more fundamental level, we may observe that the very design of a laboratory test

reflects current concepts and practice in the formulation of stress-strain-time laws, since

the tests are designed to measure the parameters contained in these laws. The ideal laboratory soil tests should (1) impose the stresses anticipated in the field on the test specimen, (2) have uniform and known stresses throughout the specimen and (3) be conducted on

test specimens truly representative of the material in situ. The tests that come closest

219
to imposing the expected field stress conditions, e.g., the simple shear box, the hollow

cylinder torsional test, and the shake table test, are not well suited to the use of undisturbed samples and do not stress the specimen uniformly throughout. We now use the cyclic

triaxial test in practice despite its nonunif ormity of stresses and the occurrence of

stress reversals which we think are not representations of actual field conditions, because

of an overriding need to get on with the job and its convenience for use with undisturbed

samples

.

Table 2 summarizes the current laboratory techniques for measurement of dynamic soil

properties and the properties obtained from each test. The shear strain amplitude

capabilities of the various tests are summarized in Figure 2.

PREDICTION/ CHARACTERIZATION OF LOAD FUNCTIONS

In seismic response problems, the load function is the earthquake ground motion, and

its prediction properly belongs to the sphere of seismology. However, the ground motions

are usually the most critical part of the input to a dynamic analysis, and it' is necessary

to specify the input motion in such a way as to have a well-posed problem for the dynamic

analysis (Idriss, 1978). In order to insure that the specified loading function (the

design earthquake) is realistic in view of the site geology and the regional seismicity,

and in addition that it is specified in a way that is appropriate for the structure

involved, the selection of the design earthquake should be a team effort, involving interaction of seismologists, geologists, and geotechnical and/or structural engineers. The

choice of a design earthquake conventionally involves the choice of such parameters of the

ground motion as the maximum acceleration, maximum velocity, and the duration of strong

motion (usually motion equal to or greater than 0.05 g). But it is generally recognized

that the specification of these parameters is not sufficient to define the characteristics

of the loading function, so one or more recorded accelerations versus time records from

some sites of similar geology, from earthquakes of about the same magnitude and epicentral

distance, and scaled to give the same peak motion parameters, are chosen as loading functions. Some of the more important issues involved in the specification of the loading

function are discussed in the following paragraphs.

220
Data Base

The United States, and the world, has experienced in the last decade a tremendous

expansion in systems of strong-motion instrumentation, and will see in the next decade or

two concomitant expansion in the available inventory of strong motion records. At the

present time, however, there are some noticeable gaps in our strong-motion data base.

A very large portion of the United States strong-motion data base consists of records

obtained during the San Fernando Earthquake of 1971. Our inventory is totally lacking in

close-in strong-motion data from earthquakes of magnitudes 7 and greater. In addition,

most of the available data represent surface motions, while it would be very desirable

to have documentation of motions occurring in the subsurface. An additional, and acutely

felt, hiatus is in strong-motion records that can be closely correlated with records of

performance of dams and other structures during past earthquakes. Such case histories

would be highly desirable for the purpose of validating our analytical methods.

Location of Input Motion

A serious question in dynamic analysis is that of where to apply the input loading

function. Commonly, the input motion is applied at the surface of the bedrock, but it is

clearly inappropriate to apply in this manner a record obtained at the ground surface. The

soil or overburden layer acts as a filter in the propagation of ground motions upward from

the bedrock. A weak layer can filter out important parts of the ground-motion, as we have

seen in strong-motion records from Niigata, Japan, where very weak motions were measured at

the surface in areas where liquefaction occurred (Seed and Idriss, 1967). Conventional

practice is to obtain transfer functions for the relations between ground motion at different levels by means of convolution through a one-dimensional wave propagation analysis.

However, one-dimensional convolution ignores lateral homogenities in the soil as well as

the decay of surface waves with depth.

Frequency Content

The input motions will not adequately test a structure if they are deficient in

frequencies in the neighborhood of the fundamental frequency of the structure, unless that

fundamental frequency is either a very high or very low one that does not normally occur in

earthquakes. Spectral representations of the ground motions such as response spectra,

221
Fourier spectra, power spectral densities, etc., while not generally used directly in

geotechnical analyses, are useful descriptors of the character of the ground-motion that

should be considered in selecting the design earthquake (Christian, 1980). Spectral

representations may also be used as the starting point in the generation of synthetic

earthquakes (Jennings, et al., 1968, Liu, 1970), which may be used as an alternative

to a natural earthquake record. A disadvantage of this is that such records are derived

from smooth spectra while the spectra of natural earthquakes are irregular, so that the

synthetic record represents too severe a motion.

SIGNIFICANT MODERN DEVELOPMENTS OR EVENTS

The Good Friday (1964) Earthquake in Alaska

The Good Friday earthquake produced spectacular damage and occurrences of ground

failures, but little useful ground motion data (Seed and Wilson, 1967). However, it produced a heightening of awareness of earthquake problems in the United States, and

instigated appropriations and research to deal with earthquake problems.

The 16 June 1964 Niigata, Japan Earthquake

The Niigata earthquake is noteworthy especially for the spectacular effects produced

by liquefaction-type ground failure. Modern structures in Niigata were seismically

designed and had adequate strength to resist shaking. However, because they were not

designed to float, a number of buildings overturned when the ground liquefied under them

(Kishida, 1966). Other evidences of liquefaction, such as sand boils, were widespread.

After the earthquake, the airport lay under a foot of water. Post-earthquake investigations in Niigata revealed useful correlations between SPT blowcounts and occrrences or

nonoccurrences of liquefaction (Koizumi, 1966, Ohsaki, 1966, 1969), gave impetus to earthquake engineering research internationally, and were the basis for liquefaction research

during the next decade.

The San Fernando Earthquake of 9 February 1971

Damage and loss of life during this earthquake were a source of widespread concern.

The slide in the Lower Sau Fernando dam during this earthquake narrowly missed becoming

the largest single disaster in United States history, since an estimated 80,000 people

222
were asleep downstream and an overtopping failure of the dam did not occur only because the

water level was below maximum pool (Seed et al., 1975). This earthquake doubled our catalog of strong-motion data, offering opportunities for the study of effects of site conditions on ground motions and empirical correlations of ground motions with performance of

structures. It is notable that there were five hydraulic-fill dams that were subjected to

accelerations of approximately 0.2 g without incurring serious damage (Seed, Makdisi and

De Alba, 1978). On the other side of the coin, the fact that so large a part of the existing data base represents a single earthquake introduces an unknown bias into the statistical characteristics of ground motions and empirical correlations. The strong-motion record

obtained at Pacoima Dam during this earthquake represented the first time that

accelerations in excess of 1.0 g have been instrumentally recorded.

Advances in Numerical Methods

The development of computer codes for one-dimensional and two-dimensional dynamic

analyses have had a far reaching influence in earthquake engineering. Studies made with

the use of these tools have provided us a better understanding of the effects of site conditions on ground motions and the relative importance of various geotechnical parameters in

influencing ground motions and soil-structure interaction (Lysmer, 1978). They have also

allowed us to exercise more realistic and sophisticated models of material and structural

behavior

.

Development of Cyclic Laboratory Tests

The cyclic triaxial , resonant column, simple shear, and shake table tests have had a

major influence on the practice of geotechnical earthquake engineering. They have given

quantitative indices of the influence of void ratio and stress level on pore pressure

development and liquefaction potential in cohesionless soils and on the influence of strain

levels on stiffness and damping (Woods, 1978). Cyclic testing methods are an indispensable

part of a rational approach to the evaluation of liquefaction problems.

Development of a Coherent Approach to Liquefaction Problems

During the 1960's and 1970' s, an approach to seismic analysis of liquefaction problems

was evolved, primarily through work at the University of California at Berkeley with

Professor H. B. Seed playing the leading role. The so-called "Seed approach" has been used

223
for the dynamic analysis for a number of major dams, and has been applied in a back,

calculation mode to the sides in the Upper and Lower San Fernando dams and to the failure

in Sheffield Dam (Seed et al., 1969; Seed et al., 1973; Seed et al., 1975).

Analysis of Post-Earthquake Pore Pressure Redistribution

Analysis of seismoscope records from the Lower San Fernando Dam indicated that the

slide actually occurred a short time after the earthquake shaking had ceased, and it was

inferred that this was due to redistribution of earthquake-induced pore pressures. Computer codes for the analysis of post-earthquake pore pressure redistribution have since

been developed (Seed, 1979b). Generally, the permeability variation within the embankment

and foundation materials are not known accurately enough to permit a high level of quantitative accuracy. However, recognition of the importance of post-earthquake pore pressure

redistribution phenomena and possible resulting instability at a later time is important

as an identification of a possible failure mechanism that had not been previously recognized. One effect this may have on the practice of seismic analysis is to place more

importance on the analysis of the effects of strong aftershocks.

Permanent Displacement Analysis

A method of analysis which treats a slide in an embankment at a rigid block on

inclined plane, subjected to earthquake accelerations, was proposed by Newmark (1965), and

a coherent analytical procedure has evolved on the basis of this concept (Goodman and Seed,

1966; Ambraseys and Sarma, 1967; Sarma, 1975; Franklin and Chang, 1977; Makdisi and Seed,

1977; and Sarma, 1979). This procedure offers a rational basis for the analysis and design

of earth and rockfill dams that do not involve materials which might be susceptible to

liquefaction. The somewhat limited experience with this procedure up to the present time

indicates that in earth dams of cohesive materials and in rockfill dams with highly permeable shells which are not susceptible to liquefaction, if there is a satisfactory static

factor of safety, direct damage, even from major earthquakes, should be limited to

relatively minor cracking or sliding.

Development of In Situ Testing and Sampling

There is at the present time a discernible trend toward greater emphasis on in situ

testing and improvement of sampling methods. There has been a great deal of study on the

224
reliability and causes of variability in the SPT test (see Geotechnical Division,

Proceedings, Specialty Conference on In Situ Measurement of Soil Properties, Raleigh, N.C.,

1975), and greater efforts to standardize the test. One notable development is the piezometer probe which measures induced pore pressures while being pushed into the soil, and

which may be useful as an indicator of liquefaction susceptibility In cohesionless soils

through indication of collapsing or dilatant behavior by the induced pore pressures

(Thorstensson, 1975; Wissa et al . , 1975; Schmertmann, 1978). Another tool under development is the dynamic pressuremeter (Mori, 1979). This equipment appears capable of measuring the in situ shear modulus over a range of strain levels and may offer the potential

for obtaining in situ values of damping over a similar range of strains. A great deal

of effort is going into refinement of geophysical testing methods, particularly in measurement of shear wave velocity. In the field of sampling of soils, we have an increasing

need for high quality undisturbed samples of cohesionless soils for laboratory tests of

liquefaction susceptibility. Until the 1970' s, the significance of structure in influencing the mechanical behavior of sands was not recognized, and it was merely assumed that a

sample of sand recompacted to in situ density in the laboratory was adequately representative of its in situ behavior. Recent research has shown that this assumption is unjustified; and that, moreover, the effects of disturbance caused by even the most careful

conventional sampling practice are very serious (Mulilis, et al .

, 197 7; Mori, Seed and

Chan, 1978, Marcuson and Franklin, 1979). A breakthrough of considerable potential importance is the experimental proof that sands may be frozen without discernible effects on

their structure, provided that free drainage is permitted away from the freezing front and

the effective stress state is maintained during freezing (Yoshimi, et al . , 1977, 1978;

Walberg, 1978; Singh, et al . , 1979). Freezing was used to aid in sampling by the Corps of

Engineers at Fort Peck Dam in 1939 (Middlebrooks , 1942) but has been little used since.

Yoshimi, et al . , (1977, 1978) report the use of radial freezing to obtain undisturbed field

samples of saturated sand, but the technique they used appears to be limited to shallow

(~ 10 ra) exploration. Development of a reasonably economical field technique that can be

used to moderate depths (~ 50 m) remains to be accomplished.

225
Solution of the Machine Vibration Problem

The conquest of this problem in the 1960's contributed a great deal to the current

state-of-the-art in analysis of soil-structure interaction problems in earthquake engineering. Of special importance is the understanding that was gained on the importance of

material and radiation damping in such problems.

Strong-Motion Instrumentation

The past decade and a half has seen a vast expansion in the number of strong-motion

instrumentation arrays throughout much of the world. The first major payoff was seen in

the United States at the time of the 1971 San Fernando earthquake when more than 100 stations produced useful strong-motion records. It is to be expected that future large earthquakes, particularly in California, will produce so many new records that they will have to

be used quite selectively (Iwan, 1978). The emphasis will likely be on filling gaps in the

data base, particularly the lack of records of nearby earthquakes of Magnitude 7 and above,

and on securing records that will document the response of particular structures to earthquakes. El Infiernillo Dam, in Mexico, was shaken by a strong earthquake (Magnitude = 7.7)

in 1979, suffering minor damage, and strong-motion records were obtained at three levels in

the embankment. Study of records of this type, together with observations of performance,

will be invaluable in validating analytical methods.

CURRENT TRENDS

From an examination of the discussions above, several trends in the development of

soil dynamics emerge. As mentioned earlier, there is a trend toward a better balance

between analytical and empirical approaches to soil dynamics problems, and particularly in

efforts to assemble and evaluate more empirical data on past performance of structures.

Advances in the various aspects of soil dynamics have been very uneven. In particular,

the development of analytical approaches, and computer codes to implement them, have

reached a level of sophistication such that the input data, rather than the analytical

models, now govern the accuracy of the analysis. Additionally, the analytical methods

need empirical validation and we are now beginning to feel the lack of well-documented

case histories. Thus, there is now a trend toward closer studies of past failures or past

performance of structures that have undergone earthquakes (Seed, et al., 1978).

226
Laboratory research to attempt to explain the cyclic mobility of medium dense sands

via exploration of differences in response during extensive and compression stress paths is

underway. This limited research done to date indicates that reversal of principal stress

directions tends to erase the beneficial effects of stress history (Tatsuoka and Ishihara,

1974).

An effective stress approach is being used for one-dimensional analysis of earthquakeinduced pore pressure development (Finn, et al., 1977, 1978). Input data are obtained from

constant volume, drained, cyclic simple shear tests. This is a step in the right direction. If the method is validated against field behavior it can circumvent the host of

problems we associate with the cyclic triaxial test and offers the potential for computing

seismically induced settlements.

The implementation of strong-motion instrumentation continues. Strong-motion

instruments have themselves been improved in recent years and we now have a better understanding of where and how to place them in order to obtain the most meaningful strongmotion data. While there are only a few major earthquake events worldwide each year, we

are now in a much improved position to learn as much as possible from most of them, and it

is to be expected that the inventory of strong-motion records will increase rapidly in the

next decade.

We are beginning to see an increased concentration of effort in the area of field

exploration, particularly the development of improved sampling methods and methods of in

situ testing of liquefaction potential and other dynamic properties of soils. Such efforts

at the present time include the refinement of the Standard Penetration Test, study and more

widespread use of the cone penetration test, the use of piezometer probes, dynamic pressuremeters, and geophysical methods, including nuclear and electrical methods directed at

measurement of the in situ density. The measurement of in situ properties by means of

seismic test methods has already seen a great deal of development and has long since

attained the status of routine test methods.

There are two areas where there is apparently little significant work in progress in

soil dynamics. One is the development of laboratory tests that meet the requirements of an

ideal test as stated earlier in Laboratory Methods section of this report. The other is in

227
the development of better undisturbed sampling methods. With the exception of in situ onedimensional freezing, we are essentially using 1940' s technology.

SUMMARY

During the last two decades remarkable progress has been made in the field of Soil

Dynamics. We consider soil response to machine vibrations to be, for all practical purposes, a solved problem. In earthquake engineering, the current state of knowledge allows

for the safe design and construction of critical structures that may be subjected to earthquake loadings; however, we are unable to evaluate the margin of safety which has been

incorporated in our current procedures. For the evaluation of the seismic stability of

existing critical structures, such as some dams and nuclear power plants, our state of

knowledge is sometimes inadequate. In the analysis of existing structures we can define

conditions which are clearly safe and conditions which are clearly unsafe. Between these

two limits lie many practical cases that fall into a grey area which will only be narrowed

by further research and new full-scale response data.

Current seismic design methodology does not rigorously account for all cause and

effect relationships. However, correction factors and compensating errors allow us to

"predict" past experience, and in this way calculational techniques have been calibrated.

More case histories are needed to further develop and refine our current approaches.

It is recognized that our numerical dynamic stress analysis capabilities are much

more advanced than our ability to obtain representative undisturbed soil samples and to

test them under the correct stress-strain conditions in the laboratory. Ongoing research

in the area of in situ testing offers us hope of being able to obtain in the future more

reliable soil properties and parameters, thus circumventing our sampling and laboratory

shortcomings

.

Finally, it is believed that the strong motion instrumentation arrays which are in

place and being expanded will provide the data which will fill the present gaps existing in

our present data base. This includes close-in records from earthquakes of Magnitude 7 and

greater

.

228
In the final analysis it should be realized that laboratory tests and analytical

calculations are done not for the sake of the numbers obtained, but for the purpose of

understanding and extending a limited field data base so that sound engineering judgments

can be made regarding the safety of structures.

ACKNOWLEDGMENTS

The authors would like to thank the WES for their support during the preparation of

this paper and their permission to publish this paper is gratefully acknowledged.

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2 32
•^GEOPHYSICAL-

—SHANNON & W

^SURFACE _ VIBRA TOR

VIBRATORY —PLATE

BEARING

STA TIC —PLATE -

BEARING

CIST —; f—SM-EO H

rj»— EARTHQUAKES

10-5 10"4 10-3 10-2 10" 1 SHEARING STRAIN-7. PERCENT

10

Fig 1 - Shearing strain amplitude capabilities

of field techniques (after Shannon & Wilson and Agbabian & Associates, 1974)

TABLE 1 FIELD TECHNIQUES FOR MEASURING DYNAMIC SOIL PROPERTIES (After Woods, 1978)

Field

technique

Refraction

P-wave S-wave Other

velocity velocity measurements Advantages

Cross-Hole

(In Situ

impulse

test)

Down-hole

(Up-hole)

Surface

SPT

Resonant

footing

Water

cannon

CIST

Depths and slopes

of layers

Velocity as function of strain

amplitude

Attenuation of

Rayleigh wave

Empirical corre- lation with

liquefaction

Modulus of near surface soils

Dynamic stiffness

of support

Constitutive Eq.

Reversible polarity

with SH-wave*

Work from surface

Samples large zone Preliminary studies

Reversible polarity

Works in limited

space

Finds low velocity

One hole

Reversible polarity

Finds low velocity

Works in limited

space

Work from surface

Widely available

Widely used in past

Work from surface

Work from surface

Pis adva ntage s Miss low velocity

zones Low strain amplitudes

Properties measured

are for thin zones near boundaries

Need 2 or more holes

Holes must be surveyed

for verticality

Needs short time interval resolution

Measure average velocities

Ambient noise near surface

Low strain amplitude

Uncertain about effective depth

Needs large vibrator

Needs "Standardization"

Limited depth of

influence

Apparatus and analysis very elaborate

Wide amplitude range Very elaborate

Horizontally polarized shear wave.

233
1

1 1—

RESONANT COLUMN (SOLID SAMPLES)

~~ RESONANT COLUMN~ThOLLOW SAMPLES)

TORSIONAL SHEAR (HOLLOW SAMPLES)

PULSE METHODS CYCLIC TRIAXIAL

CYCLIC SIMPLE SHEAR

SHAKE TABLE

TYPICAL MOTION CHARACTERISTICS

PROPERLY DESIGNED STRONG GROUND CLOSE-IN MACHINE FOUNDATION SHAKING FROM NUCLEAR

|

| EARTHQUAKE | EXPLOSION

10"4 10-3 10-2 ICH 1 SHEARING STRAIN AMPLITUDE (%l

Figure 2. Shearing strain amplitude capabilities of

laboratory apparatus (after Woods, 1978).

TABLE 2

LABORATORY TECHNIQUES FOR MEASURING DYNAMIC

SOIL PROPERTIES (After Woods, 1978)

Shear

modulus

Resonant

column

with

adaptation

Ultrasonic

pulse

Cyclic

triaxial

Cyclic

simple

shear

Cyclic

torsional

shear

Shake

table

Young '

s

modulus

Material

damping

Cyclic Stress

behavior Attenuation

234
DYNAMIC ANALYSIS OF SHEET PILE FOUNDATIONS

by

Tatsuo Asama

Yukitake Shioi

Michio Okahara

Yasuo Mitsuie

ABSTRACT

The sheet pile foundation is a new type foundation with steel pipe sheet piles placed

in the form of a circular, or oval well to produce behavior intermediate to that of a

caisson and a pile.

For a reasonable design of this foundation type, it is necessary to fully consider the

effect of shearing deformation of the joints between the sheet piles. From a past Investigation, a program for stability calculation using a seismic coefficient method makes possible the consideration of the joint effect. By utilizing this program, a dynamic analysis

program permitting time hysteresis response calculation was developed.

Using such a dynamic analysis program, a case study which examines the dynamic behavior

of a sheet pile foundation was completed.

235
INTRODUCTION

The sheet pile foundation is a new type foundation with steel pipe sheet piles in the

form of a circular, or oval well to produce behavior similar to that of a caisson from the

bonding effect of joints between the sheet piles, and the restraining effect of the footing

at the head of the piles. The outline was reported at the 7th meeting of the UJNR Wind and

Seismic Effects Panel. [1] The authors continued to examine an earthquake resistant design

calculation method which would permit evaluation of the structural characteristics of this

foundation. They were able to obtain good results with respect to a static design

calculation method by using a seismic coefficient method. [2]

Then, from the results of the static design calculation method investigation, the

authors developed a calculation program which would permit the time-hysteresis response

calculation of the sheet pile foundation. They examined the dynamic response characteristics of the sheet pile foundation, and the result of this examination is reported here.

SEISMIC RESPONSE CALCULATION PROGRAM

To examine the seismic response of a structure embedded in the ground such as a sheet

pile foundation, it is necessary to consider not only the effect of the seismic movement of

the bearing layer at the tip end of the foundation, but also the vibration of the subsoil

on the sides of the foundation.

To incorporate the so-called interaction of the foundation and the subsoil, it is a

general practice to represent the foundation and subsoil by models using the finite element

method, or to replace it with a concentrated mass point - spring system. In this program,

the latter method was employed in order that a solution with a relatively small number of

degrees of freedom would be obtained.

The vibration model is shown in Figure 1. It is composed of a foundation model and a

subsoil model. These are connected to each other using horizontal springs, vertical

friction springs, and rotary springs.

In the foundation model, the element cut crosswise in the direction of depth of the

sheet pile foundation was substituted by the mass points, and it was assumed that the mass

236
points were connected with one another across the springs and dash pots. To calculate the

stiffness matrix in the vibration system of the foundation, the model was designed to

permit consideration of the shearing deformation of the joints between the sheet piles.

In the response calculation, the response of the subsoil only was first calculated

from an earthquake input to the bearing layer, then the response of the foundation model

was calculated with the subsoil vibration thus obtained then taken as an input seismic

wave to the foundation model, and the response values of displacement, acceleration,

moment, and axial force were obtained.

Figure 2 shows a flow chart of the program.

RESPONSE CALCULATION

The structural characteristics of sheet pile foundations are next examined using the

dynamic analysis program.

Calculation Conditions

Through a series of examinations using a static program which employs the seismic

coefficient method, the parameters which have a large influence on the response are

extracted, rigidity Gj to the shearing deformation of the joints between the sheet piles,

and the ratio of the frictional spring constant Kf between pile and subsoil to the horizontal spring constant or Kf/Ky. The following case study will examine the effects of

these parameters.

In calculations, the following conditions are assumed.

Sheet pile foundation: Diameter, 15 m, Single sheet structure

Sheet pile: Diameter, 1000 mm; Thickness, 15 m; Steel pipe pile

Joint: Diameter, 165 mm; Thickness, 11 mm; Steel pipe

Embedded depth: 60 m, 30 m, and 20 m

Subsoil conditions: Tip end (2.5 m) N value 50

Peripheral surface, N value 5

Calculation cases are shown in Table 1. In the table, Gjq represents the joint

rigidity obtained by experiment [1], and is taken as 120,000 t/m^.

237
The values of the horizontal spring constant, Ky, and the bottom spring constant, Ky,

were calculated by conversion from the N values following the Specifications for Highway

Bridges

.

For handling the subsoil the circular form sheet piles, all the soil was considered

as added mass.

For the input seismic wave, an artifical seismic wave having as flat a frequency

characteristic as practicable was provided so that the frequency characteristics of the

vibration system would least affect the response. It was input at a time interval of 0.02

second with a maximum acceleration at 100 gal (Figure 3).

Calculation Conditions

As an example of response, a stress distribution in the direction of depth with

Kf/Kg =0.5 is shown in Figure 4. This is calculated using the pile producing the maximum

stresses in the direction of vibration. As seen, a very great stress is produced close to

the pile head. A similar phenomenon is observed in the static analysis; but, in the

dynamic analysis, a fluctuation of the stress close to the pile head is characteristically very sharp. The stress close to the pile head is due mainly to the pile head moment.

However, as the joint rigidity Gj increases, the pile head moment decreases, resulting

in a decrease of the stresses. Such effect of Gj is similarly produced in the static

analysis

.

The Gj effect was examined as to displacement, acceleration, and cross-sectional force

as shown in Figure 5. The response values at the pile head of acceleration, axial force,

and bending moment, but not displacement, decrease with increasing Gj. Further, such

decrease is appreciable at the stage of Gj/Gjq = 0.1. Thus, the importance of including

the effect of Gj in design calculations is confirmed. Additionally, that the pile end

axial force increases with increasing Gj indicates that the sheet pile foundation changes

its behavior from that of the pile foundation to that of the caisson. Thus, if Gj is not

taken into consideration, the design is apt to be on the unsafe side.

Such trends were also confirmed in the static analysis.

Figure 6 shows the result of an examination of the effects of the frictional spring

Kf which was changed by 0, U.2, 0.5 and 1.0 times of the horizontal spring with Gj

fixed at 0.1 Gj(J.

238
The sheet pile foundation is executed as piles so that the proportion of the

circumferential frictional force to the total bearing force is considered to be fairly

great. In this case, it is confirmed from the result of calculations according to the

method of calculation of the cirmcumf erential frictional force of pile foundation that

Kf = 0.5 Kjj is a reasonable value.

With increasing Kf , pile head displacement, acceleration, and axial force decreases,

while the bending moment tended to increase. That the pile end axial force can be reduced

by considering Kf is a very advantageous trend for the design.

SUMMARY

As the result of time hysteresis response analysis of the sheet pile foundations by

input of artificial seismic waves, the following is confirmed.

(1) A great stress concentration occurs close to the pile head.

(2) As joint rigidity, Gj, between the sheet piles increases, the pile head response

decrease, but the pile tip axial force increases.

(3) As the frictional spring, Kf , increases, the pile head bending moment increases,

but the pile tip axial force decreases.

(4) The foregoing results agree well with the results confirmed by the static

analysis

.

REFERENCES

[1] Kawakami, K. , et al., "Sheet Pile Foundation and its Structural Characteristics

Against Horizontal Loads," 7th Joint Meeting, U.S. -Japan Panel on Wind and Seismic

Effects, May 20-23, 1975.

[2] Shioi, Y., et al., "Structural Analysis of Sheet Pile Foundation," Preprint of

25th Symposium on Structural Engineering, February 1979.

239
Z50

12 -^-s

Legend

/ —\ Springs and mass

^-rj^j— of subsoil model

(T) Rotary mass of the top slab

B Horizontal mass

of the foundation

Fig. 1 Vibration model

Table 1 Calculation cases

A B C D/L 15/60=0.25 15/30=0.50 15/20=0.75

kT/k^Gj/Gjo 0. ! 0.1 0.5

i -j

1 .0 0. 0.1 0.5 1.0 0. 0.1 0.5 1.0

1

0 . o

' o j o —H \ o o ] o 0 0 0 0

1 O ; O 0.2 0 I 0

1 o

' 0.5

1 1 O 0 i o

. o O 0 0 o 0 o O O 1.0 1

i

1 1

1 ° o

1

240
Foundation model

Stiffness matrix (Support

displacement considered)

Mass matrix

Subsoil shearing model

Stiffness matrix

Mass matrix

Characteristic value and mode calculation

Characteristic value and mode calculation

Foundation model and subsoil mode mode

combination

Response calculation

(Modal analysis)

1 Response waveform

output

Input of ground movement at input

boundary

Fig. 2 Flow chart of dynamic program

TIME«0.0 - 15.0 [SEC] DT-0.0J

MAX. 99.37 (GAJ-J

g

4

POWER - SPECTRUM

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 11) 17 IB 19 20 21 22 23 24 25 FREQUENCY (Hz) Fig. 3 Artificial earthquake waveform

and power spectrum

241
MODEL-B

Pile stress intensity (kg/cm2

) 4 Distribution of the maximum pile stress intensity

in the direction of depth (Kf/KH =0.5. W=100%)

Fig. 6 Effects ot Frictional spring (Gj=12 ,000t/m 2

; inside

soil weight, 100%)

242
DYNAMIC BEHAVIOR OF MULTI-COLUMN FOUNDATIONS

WITH INCLINED PILES DURING EARTHQUAKES

by

Tatsuo Asama

Yukitake Shioi

Tasuo Mitsuie

ABSTRACT

To construct a pile foundation in a lake or sea, a novel method of placing piles from

the water surface without coffering and constructing footings in air has been used

recently. Here, the piles work as columns and is thus called the multi-column foundation.

This multi-column foundation is distinguished in workability but is lacking in

rigidity to resist horizontal external forces. Particularly, the conventional multi-column

foundations are composed of only vertical piles so that with an increasing free length of

columns, a greater number of piles are required in order to keep the horizontal displacement below the allowable value.

In such a case, the joint use of inclined piles in a multi-column foundation is

desirable. However, the multi-column foundation with inclined piles has no precedent,

and its structural characteristics required that many points be resolved. In this paper,

the dynamic behavior of the multi-column foundation using inclined piles is examined by

calculating the time hysteresis responses.

243
INTRODUCTION

The multi-column foundation is a type of foundation developed for improving the

workability of pile foundations in a lake or sea, and its use is increasing.

However, where the free length of pile is long and the subsoil is relatively soft,

the pile has its horizontal resistance reduced, and in order to keep the displacement of

the footing within allowable values, numerous piles are required.

Thus, in order to enhance the rigidity of the foundation, the use of inclined piles

may be considered. However, the structural characteristic of the multi-column foundation

has many points which need clarification.

This investigation is intended to clarify the structural characteristic of a

multi-column foundation comprising of inclined piles, and thus the effects of inclined piles

introduced upon the response displacement, acceleration, and cross-section force through

calculation of the time hysteresis responses of the pier by using actual seismic records

as input data both with and without the inclined piles.

METHOD OF INVESTIGATION

Vibration Model

(1) Foundation Model

A multi-column foundation of the following dimensions is assumed.

Number of piles: Longitudinal direction, 7 rows;

Transversal direction, 9 rows; a total of 63 piles.

Where inclined piles are included, the piles along the outermost

row are inclined at 8 =15°.

Piles: Steel pipe pile, $ 2500 mm, t 40 mm;

Cross-sectional secondary moment I = 0.234 m

4

;

Area A = 0.309 m

2

; Free length h = 38 m; Embedded depth D

f = 37 m

Footing: Pile interval, 2.0 (f

LG direction 35 m x TR direction 45 m x thickness 3.5 m

LG direction I = 20.7 m4

, A = 20.3 m

2

TR direction I = 16.1 m

4

, A = 15.8 m2

E

c = 2.7 x 10 6 t/m2

244
Body of pier Height, 22.8 m

LG direction, I = 5.40 m4 , A = 3.28 m2

TR direction, I = 139 ra 4

, A = 3.28 m2:

E

c = 2.7 x 10 6 t/m2

The multi-column foundation having such dimensions is replaced by a plane rigid frame.

In this case, the cross-sectional dimensions are so taken that in the rigid frame of the LG

direction they are respectively a multiple of the number of rows in the TR direction, and

in the TR direction they are a multiple of the number of rows in LG direction. As concentrated masses, a plane rigid frame structure supported by concentrated springs is assumed,

as shown in Figure 1.

(2) Subsoil Model

It is assumed that the subsoil is composed of three layers, with the modulus of

deformation of the respective layers taken as shown below.

Upper layer (Ground surface ~ -10 m) 10 kg/cm2

Middle layer (-10 m 20 m) 50 kg/cm2

Lower layer (-20 m ~ ) 100 kg/cm2

From these values of the modulus of deformation, the horizontal subsoil reaction

coefficients were obtained according to the Specifications for Highway Bridges, and thus

the subsoil was modeled as concentrated springs.

Method of Response Calculation

The time hysteresis responses, using actual seismic records as input data, were

calculated

.

As an input seismic wave, the records of the Izu Peninsula South Coast Earthquake

(May 1974) and the Hachijo Island East Sea Earthquake (December 1972) observed in Ukishima,

reclaimed land in the Tokyo Bay, were used. The records include acceleration observed at

depths of 0 m, -27 m and -67 m. Figure 2 shows the records at ground surface. It will be

seen that the waveform of the Izu Peninsula South Coast Earthquake is predominantly in the

long frequency component, more so than the Hachijo Island East Sea Earthquake.

In the response calculations, the record at 0 m was used as input simultaneously to

the joint at the ground surface, the record at -27 m to the joint at -19 m, and the record

245
at -67 m to the joint at -37 m. The response calculation was executed with the maximum

acceleration taken as 100 gal (at ground surface), time interval at 0.02 sec, and duration

at 25 sec ~ 30 sec.

For the damping constants, 10 percent was taken for the piles in the ground and

2 percent of the piles on the ground surface, the footing, and the body of the pier.

RESULT OF INVESTIGATION

Modes of Vibration

Table 2 shows the natural periods of the respective modes in the case of both straight

and inclined piles. Figures 3 through 6 show the primary to the fourth modes of vibration.

In both cases of straight and inclined piles, the foundation Is a structure of very

long period. However, in the example of the primary mode, it will be seen that the natural

period is reduced considerably by the use of inclined piles. In the secondary and subsequent modes, partial vibration is prevailing so that it seems that there is little effect

from the introduction of the inclined piles.

Response Displacement

Figure 7 shows the time hysteresis responses of the footing displacement using the Izu

Peninsula South Coast Earthquake input. In the case of the straight piles, the response

displacement is very great, and a gradual increase of the response amplitude is noted.

Where the inclined piles are used, however, the maximum displacement is smaller by about

the order of one, and the displacement peaks at about 15 sec and decreases thereafter.

Table 3 lists the maximum displacements at the respective joints. The maximum

displacement of the footing is different by about an order of one due to the characteristics of the seismic waves. It is also seen that the displacement decreases considerably

when using the inclined piles.

Axial Force

Figures 8 and 9 show the maximum axial forces produced in the respective members.

It will be seen that where an inclined pile is included, a great axial force is produced in the inclined pile. However, the axial force produced in the inclined pile is by

246
no means of a great value when compared with the maximum axial force of the straight pile

in the case of straight piles only. This seems to be due to a decreasing maximum response

displacement when using the inclined piles.

Bending Moment

Figures 10 and 11 show the maximum bending moments in the respective members. It will

be noted that the bending moment at the pile head or at the lower end of the body of the

pier is reduced greatly by use of the inclined piles.

SUMMARY

Table 4 lists the maximum values of the response.

The results of the investigation are summarized below.

(1) By introducing the inclined piles, the natural period in the primary mode of

vibration is reduced considerably.

(2) Using only straight piles a phenomenon of the response increasing gradually

depending on the input seismic wave was observed.

(3) The response values as displacements, axial forces, and bending moments are

reduced considerably by introducing the inclined piles.

From the foregoing, it is confirmed using dynamic analysis that inclined piles are

very effective in the case of soft structure foundation types, such as the multi-column

foundation

.

Thus, in the case of a multi-column foundation with great free length, it is necessary

that the foundation itself be given a rigidity to some extent, and for such purpose, the

introduction of inclined piles is desirable where the work, conditions permit.

247
Table 1 Cross-sectional dimensions

(1) Straight pile* only, Ill Table 1 Cross-sectional dimensions

jl dire., t i en I »> StrjiB ht pi 1

Member Attn'

)

Km*) Member, hum. No. Conner! ion

1 278 2 1 1 .'. 1 10' U. 1U

l-« - IS- J12-29- (6-4 J- SO

I-10-17-J4-3I- JM-i 5-->.' 1-18-23-3;- 19-46-53 5-12-19-26-J 3-40-47-54

b-l 1-20-27-14-41-48-55 7-14-21-28-35-42-49-56

2 2.78 2.11 1 . 1 1 U

' U.II2

50- 57-64-71-78

51-58-65-72-79

52- 59-66-7 1-80

5 1-60-67-74-81

54-61-68-75-82

55-62-69-76-8 i 56-61-70-77-84

3 -•[>. ) JO. 7 -.110 /8- 79-80-81-82-8 i-8»

4 3.28 5.40 . 1 1 u

81-85-88

Member, Joint, No. Connection

1- 10-19-28-37-46-55-64

2- 11-20-29-38-47-56-65

-12-21-30-39-4 8-57-66 -13-22-31-40-49-58-67 -14-23-32-41-50-59-68

6- 15-24-33-42-51-60-69

7- 16-25-34-4 3-52-61-70

8- 17-28-35-44-53-62-71 -18-27-36-45-54-63-72

64-73-82-91-100

7i-83-92-101 86-75-84-93-102

67- 76-85-94-103

68- 77-86-95-104 by-78-87-96-105

70- 79-88-97-106

71- 80-89-98-107 -81-90-99-108

100-101-102-103-104-105-106-107-108

111 ll pile. Iuiiri < lid i

Member Aim- 1 Km* 1 11 I /m ) l).iin|iitiy Kji iu

1 278

1

2.11 . 1 • 1 u

0. ID

l-H -I5-JJ-J9- ih-'. »_q - | h- l- )0- P-ltU- 1

, 1 J-lu-l ?-j4- |I-3rt-4VSJ ft-ll-lH-iS-)i-W-«6-Si

)>- 1 1- 'U-.'/- (4-4 1 -iH-SS 7-I4-LM-J8- )V4.'-49-S6

2.78 2.11 1. 1 • 10

'

in- 37-h ,- / 1 - /k ,l_SM-h*)-7j-74 Sj-^y-bh- 7 i-KIl

S l-«l-*t7-?4-KI )4-6l-bt!-7*-HJ iS-hJ-hV- 7C-M 1 '>0-bJ-7U-7 7-h-.

3 20. i 20.7 J . 1 1 u

7tt-79-8U-ttl-H2-ff

3.28 5.40 2 I

- J u

1 « 1 -H*>-«f-.

Member, Joint, No. Con

10-19-28-37-46-55-64

2- 1 1-20-29-38-47-56-65

3- 12-21- 30-39-4 8-57-66

4- 1 1-22-31-40-49-58-67

5- 14-23- 12-41-50-59-68

6- 1 5-24-13-42-51-60-69

7- 18-25- 14 -4 3-52-61-70

8- 1 7-28- J5-44-53-62-71

9- 18-27-36-45-54-63-72

64- 7 1-82-9 1-100 65- 74-81-92-101 68-75-84-93-102

87- 76-85-94-103

88- 77-88-95-104

89- 78-87-96-105

70- 79-88-97-106

71- 80-89-98-107

72- 81-90-99-108

100- 101-102-103-104-105-106-107-108

04-109-110

243
Table 2 Natural periods of the modes (second;

\. Models

Mode \.

orders ^s.

Straight piles only Inclined piles included

Longitudinal

direction

Transversal

direction

Longitudinal

direction

Transversal

direction

1 2.44 2.38 1.40 1.40

2 0.90 0.39 0.90 0.39

3 0.38 0.38 0.39 0.39

4 0.37 0.37 0.38 0.38

5 0.36 0.36 0.36 0.37

6 0.36 0.36 0.36 0.36

7 0.36 0.36 0.36 0.36

8 0.36 0.36 0.36 0.36

9 0.33 0.36 0.32 0.36

10 0.19 0.36 0.18 0.35

Table 3 Maximum displacements (cm)

Longitudinal

direction

Transversal

direction

Location Seismic wave

Straight

piles only

With

inclined

piles

Straight

piles only

With

inclined

piles

Footing

Hachijo Island 9.22 1.78 9.45 2.35 Izu Peninsula

South

61.22 6.74 89.33 8.90

Midpoint of piles on the ground

surface

Hachijo Island 5.74 1.31 5.96 1.70 Izu Peninsula

South

38.51 4.92 56.89 7.00

Piles at groundd

surface

Hachijo Island 1.52 0.41 1.60 0.52

Izu Peninsula

South

10.28 1.45 15.31 1.87

Midpoint of piles in the ground

Hachijo Island 0.03 0.02 0.03 0.02

Izu Peninsula

South

0.15 0.04 0.24 0.05

249
Table A List of responses

Shown dotted is the cross-section in

N,,ted poinl Input wave

A., eleratlon (gal) Horizontal dis placement (cm) Longitudinal Longitudinal Transversal transversal direction St ra.ghl piles only

piles included

Straight

piles only

In, lined

pi les

ini luded

Straight

piles only

Ini lined

pi 1 es

inc 1 uded

Straight

piles only

Inc lined

Included

u Pooling 1 lai h i jo 1 s 1 a nil 7 7 5b 1 2 1 65 9 . 2 2 1.78 9.45 2.35

i

i

Izu Peninsula 145 69b 186 bl.ll 6.74 89.33 8.90 Midpoint "1 u

piles on

the ground

liai hi jo 1 slund 69 1 111 84 5.74 1.31 5.96 1.70

; -4—

1

1

1

Iztl I'ii sola 249 1 26 444 1 68 38.51 4.92 56.89 7 .00

Pi les at

1 ground

ll.lilli jo Island 51) 57 1 .52 0.41 l.bO 0.52

i —I1 l/u Peninsula 7h 1 1 64 10.28 1 .45 1 11 1.87

5

1

Midpoint nf

- piles in the ground

Ha. hi jo Island If) l« 0.0J 0 .02 0.03 0.02

Ixu Peninsula 19 5 0.15 0.04 0.24 0.05

Noted pi lint Input wave

Axial l.irt:* ( t 1 Per pi 1e Bend i tig moment ( t .tit) Per pile Shearing force (t) Per pile

hoiin Uittlin 1

1

1 ong i t nil i na 1 longitudinal Transversal

Straight

piles only

In. 1 i nod

pi les

in. ludod

straight

piles only

In. 1 i noil p i 1 es

in. ludod

Straight

piles nnl>

1 ii, 1 i lied p, les

in. lu. led

straight Piles only pi les

in, lulled

Straight

piles only

1 nc 1 ined piles

i in 1 uded

Straight

piles only

I nc 1 ined piles included

Foot ing

Hui hi jo Island 14 1

' IKU • 127 -".1 77K 2 ill Kill 2b4 2 5 8 28 9

•U Peninsula '941 •666 1 I ill 9 1 I <H78 7hl 7 2 24 984 154 14 223 18 Midpoint of piles on

the ground surface

Haihijn Island '14

)

• IHI .14 1

1 5h 1(17 1 79 1 J 1 32 10 34 12 Uu Peninsula • hhK •1 141 4 1 (. 971 1 (I'll 1 54 1 409 204 35 304 46

Kile;, dt

g round surface

Haihl jo Island ' 1 '. 4 1 M 1 1 28 a i 4 7

.'

1 1 1 144 12 13 36 17 I/u Peninsula 94 1 h(>9 '114 1 91 7 40 1 5J7 40 .12 1 53

Midpoint of piles in the groundHa, hi jo 1 sland ' 1 iS .'II II

1

4h 1 14 1 28 J

1

9 1 1 12 Uu Peninsula '94 1 587 1 14 i )J4 11129 4 ) 1 20 1 30 29 7 41

78 79 80 81

43'*M4l/M5-1*461*47 yW8>W49(>(f — - ~,o —

o

36 *V37 W38 VA39 Vr40 W41 VV*2 » -— K> o

29 W30 ",31

22< W?y *4 >W2S rW26 W27 W-28 W - "

15M"16^17mi8*19*?0Wl21 W JStl? _

cc Btt"l»IO#ll#l!|tllJ|fH« -*-H ' IW—

l<ktf2W3fW4«W5<M, 6<rrf 7 (*

W« 32^33 H134 Vj»35 *» * H-« g

6.5.000=30.000

. Joint I Hh O Mass point w Sprino

101 10? 103,

37 W3R »>39 W40 W*41

10 (IMl W12 W13 W14 fl5Hll6«117

105,106,107 108.

55( >*56r *5 7( «5Rr W59< >*6(y >*6 V >W6?: >W63( >»

46 »47 »48 Vr49 Vr50 *>51 W52 *>53 W54 (*

28( *29, IH3W >*ni' >W32 *B3< *34- >*35r l*36( * — —

19 1*20 **21 *"2? W-23 W"24 W25VP26 *«27 W

* H2

W 5 b# 6 *l 7, A 4 i * 4

W42 #43 *»44 t*45 *>

#8 #9'i«l

8 .5.000 40 000

• Joint IM 10 O Mass point # Spring

(1) Straight piles only, longitudinal direction (2) Straight piles only, transversal direction

Fig. 1 Dynamic models

250
Fig. 1 Dynamic models

79 80 81

52

741) 75 0 76 " 77

67 ) 68<> 69"

53

KH6' 44 0W450W46(>#17I>*48<>»KH6

37 #38 #39 #40 #41 #KH5

30 tf- 31 #32 #33 # 34 #KH4

3 $»>24 ( «t'2s /W26< l#27< * K H 3 #18 #19 #20 #KH2

9 |#10 #11 #12 #13 #KH!

20# 3<>#4<>*>6<>#6<>#KS

4»5.OOO = 20.0C0

55

• Joint I Sb O Mass point # Spring

Inclined piles included, longitudinal direction

• Joint 1110

o M;iss point Spr i nj>

(A) Inclined piles included, transversal direction

251
Ukishima, Kawasaki, Kanagawa (NS component)

Seismic center - Izu peninsula south coast

0.00 8.00 16.00 24.00 32.00 40.00 48.00 56.00 64.00 72.00 80.00 TIMF

(2) GR. UKISHIMA N-S

Fig. 2 Seismic waves used in analysis

(Records of acceleration on the ground surface)

252
253
T=0.90 T=0.39

T=0.91

T=0.40

Fig. 4 Mode diagrams (Secondary)

2 54
T=0.38 T=0.38

Fig. 5 Mode diagrams (Third)

255
Fig. 6 Mode diagrams (Forth)

256
s + + + + + +tl.5

0 0.5 1.0 1.5 2.0 2.5 3.0 EI TIME (SEC) X10

(b) Inclined piles included

Fig. 7 Seismic responses

(Footing; Izu peninsula south coast earthquake)

257
Longitudinal direction,

straight piles only Axial force diagram (t)

coast earthquake

-Hachi] 0 Island eaac sea earthquake

Fig. 8 Axial force distribution diagram

Transversal direct ion, inclined piles included

Axial force diagram (t)

llu peninsula south

coast earthquake Hachijo island east earthquake

Fig. 9 Axial force distribution diagram

258
Fig. 11 Bending moment distribution diagram

259
ANALYSIS OF LIQUEFACTION POTENTIAL OF SANDS ON LEVEL GROUND

DURING EARTHQUAKES USING CYCLIC STRAIN CONTROLLED TESTS

by

Felix Y. Yokel

Riley M. Chung

Ricardo Dobry

t Richard S. Ladd

INTRODUCTION

The current state-of-the-art method of predicting liquefaction potential of saturated

level sandy sites during earthquakes was recently summarized by Seed (1979 [1])- In this

t method, the cyclic stress ratio ( T /°v

) e

is compared with the cyclic strength ratio

i (t/o

v) cs where:

t = cyclic shear stress

a y

= effective vertical overburden pressure

(T/o v) e = cyclic stress ratio caused by the earthquake

( T /° v) cs

= cyclic stress ratio needed to cause liquefaction

(cyclic strength ratio)

i t Whenever ( T /° v ) e

> ( T/, °v^cs liquefaction is predicted.

t Two procedures have evolved for obtaining the cyclic strength ratio ( T /° v

) cs

- The

first procedures relies on empirical correlations between the Standard Penetration Test

(SPT) results and liquefaction potential derived from information from sites where liquefaction occurred in previous earthquakes. In the second procedure the cyclic strength

ratio is determined by stress-controlled cyclic triaxial tests. Both procedures have problems as discussed by Seed (1979 [1]), Peck (1979 [2]) and Dobry et al. (1982 [3]). Some of

the problems are discussed below:

While it has been demonstrated that there is a correlation between the SPT blow count

and the liquefaction potential of saturated sands, this correlation has considerable scatter and provides definite answers only for extreme situations. The SPT test itself is not

sufficiently standardized and the results obtained vary with the equipment and procedures

260
used to perform the test [i.e., see Kovacs (1979 [4])]. This variation raises serious

questions about the interpretation of historical SPT data for which there is no record of

equipment and procedures used

.

The laboratory determination of the cyclic strength ratio is the more unreliable of

the two methods. The parameters which affect the cyclic strength ratio, such as soil

fabric, overconsolidation, geological age, and seismic history, cannot usually be determined for the site of interest and thus are not simulated in the laboratory. Neither is

it possible to retrieve truly undisturbed samples of sand for laboratory testing. Peck

(1979 [2]) discussed the laboratory procedures and concluded that they lead to decisions

which are unnecessarily conservative and expensive.

The approach recommended in this paper is to predict liquefaction potential on the

basis of predicted cyclic strain, rather than cyclic stress, and is designed to eliminate

some of the difficulties encountered in the previously discussed state-of-the-art methods.

EFFECT OF SOIL FABRIC ON PORE PRESSURE BUILDUP

The effect of sample fabric on the cyclic strength ratio is illustrated by the cyclic

triaxial test results in Figure 1. In this figure the cyclic stress ratio is defined by as

• i t/o-j where a-j is the effective confining pressure prior to application of the cyclic load.

The circles represent data from tests performed by Park and Silver (1975 [5]) and discussed

later by Yokel et al . , (1980 [6]). Note that at a given stress ratio, specimens prepared by

wet rodding will resist 15 to 20 times the number of stress cycles producing failure in

specimens of the same relative density prepared by dry vibration. Figure 2 shows the

buildup of pore pressures as a function of the cyclic stress ratio and the number of stress

cycles for the same stress-controlled tests shown in Figure 1. It can be seen from

Figure 2 that in this type of test the effect of sample fabric on pore pressure buildup is

more important than the effect of the number of test cycles applied. This is an important

(and discouraging) conclusion. The number of stress cycles is related to the duration of

the ground motion and can be estimated. However, soil fabric cannot be presently (1980)

measured in situ and thus its effect cannot be evaluated in an engineering study.

261
RELATIONSHIP BETWEEN CYCLIC SHEAR STRAIN AND PORE PRESSURE BUILDUP

General Trend

Figure 3 shows a curve originally proposed by Dobry and Ladd (1980 [7]) for the pore

pressure buildup, Au, during strain-controlled cyclic triaxial tests on sands having a

relative density (Dr ) of approximately 60 percent. Figure 3 is for 10 cycles of strain;

shear strain was taken as 1.5 times the cyclic axial strain. The first three sands identified were tested by Park and Silver (1975 [5]). The other two sands are from test data

obtained by Dobry and Ladd (1980 [7]). More recent tests performed by the authors on

Monterey #0 sand also closely follow the solid curve in the figure which shows the trend

of the test data.

It can be seen that the curve in Figure 3 very well represents data from different

sands (crystal silica #20 sands, pit-run sands, and Monterey #0 sands), three methods of

i sample preparation (fabric) and a considerable range of confining pressures (o

c

) . Also,

the data in Figure 3 were obtained independently at two different laboratories. This evidence strongly suggests that the curve in Figure 3 is generally applicable to most sands

with a relative density of about 60 percent.

One of the most significant features of Figure 3 is that up to a shear strain of

approximately 10~2 percent there is essentially no pore pressure buildup.

The trends shown in Figure 3 were further investigated in a program of cyclic triaxial

strain-controlled tests, conducted by the authors on saturated Monterey #0 sand. Some of

the results of these tests are discussed hereafter. More detailed information is presented

in the final report on this investigation by Dobry et al., (1980, issued July 1982 [3]).

Threshold Strain

Figure 4 shows a plot of residual excess pore pressure, u, versus cyclic shear strain,

Y, up to 3*10~2 percent. These results are discussed in some detail by Yokel et al . (1980

[6]). The vertical scale which shows the residual pore pressure ratio is deliberately magnified to show even a minor increase of pore pressure. Note that below the cyclic shear

strain of approximately 10~2 percent there is no pore pressure buildup. This confirms the

observation made in Figure 3 and suggests the existence of a threshold strain, y*. ~ 10

percent. Figure 5 shows the effect of confining pressure on Yt f°r tne range 500 psf

262
(24 kPa) < o'

c

< 4000 psf (192 kPa) . Figures 4 and 5 indicate that the value of Y t is quite

i insensitive to both, relative density (D

r

) and confining pressure (o

£ ). These figures also

show that up to a cyclic shear strain y - 3«10-

^ percent the pore pressure builtup after

10 cycles is very small and is essentially a unique function of y.

The lack of pore pressure buildup for y <^ y t can be explained by the loading deformation behavior at the sand grain contacts at very small cyclic strains. This load-deformation behavior was originally investigated by Mindlin and Deresiewicz (1952 [8]) who studied

two elastic spheres in contact pressed together by a normal static load and subjected to a

tangential cyclic load. These authors derived an expression for the tangential displacement which can be taken as a measure of strain. Dobry and Swiger (1979 [9]) applied this

solution to calculate threshold strain (Y t

) as a function of confining pressure (o" c) for

a cubic array of quartz spheres using the elastic constants and friction coefficient typical for quartz. Threshold strain was taken as the strain at which gross sliding between

the spheres is imminent. Below this strain the sliding between spheres and the resultant

volume change are frictionally blocked. The result of this work is shown in Figure 6.

Note that within the range of confining pressures which is of more practical interest

[500 psf (20 kPa) to 4000 psf (200 kPa)] the threshold strain does not vary significantly

and is reasonably consistent with the experimentally-observed threshold strain.

The existence of the threshold strain and its practically constant value are of great

practical significance, since sites which are subjected to seismic loads and experience

strains smaller than the threshold strain will not liquefy, regardless of the number of

applied load cycles. For small strains comparable to Yt tne anticipated shear strain can

be calculated if the maximum shear modulus of sand is known. The maximum shear modulus,

in turn, can be reliably determined in the field by geophysical measurements. Therefore,

a design criterion based on Yt can t> e developed. Such a criterion is suggested by Dobry

et al . (1980 [10]). It is especially useful when considering a stiff sand site subjected

to earthquake accelerations up to 0.3 g.

Effect of Number of Strain Cycles on the Pore Pressure Ratio

The effect of cyclic strain and of the number of strain cycles on Au is shown in

Figure 7 for a wide range of cyclic strains. It can be seen from this figure that up to

100 strain cycles there is no pore pressure buildup below the threshold strain. It is

263
also of interest to note that hu/a^ is below 0.2 up to a strain of 3*10 percent even

after 100 cycles. Thus it is reasonable to conclude that at the strain of 3*10~2 the pore

pressure ratio during an earthquake would not exceed 20 percent.

Effect of Relative Density on the Pore Pressure Ratio

The effect of relative density (D r ) is shown in Figure 8. In the tests shown in this

figure the confining pressure is held constant at 2000 psf (95.8 kPa) and the number of

cycles at 10. The range of relative densities is from 45 to 80 percent. Figure 8 shows

that Dr is a significant variable in determining pore pressure buildup for cyclic shear

_o strain y > 3*1- percent and that the importance of D

r increases as y increases.

CONCLUSIONS

Several important conclusions can be drawn from the data presented in this paper:

(1) Saturated sands will not experience pore pressure buildup when subjected to a

cyclic strain smaller than 10"^ percent, regardless of the type of sand involved, its

relative density, its fabric, and the number of strain cycles.

(2) When subjected to up to 100 cycles of strain smaller than 3»10~2 percent, the

pore pressure ratio in saturated sands will probably not exceed 20 percent, regardless of

the relative density of the soil.

(3) There is strong evidence indicating that the effects of soil fabric on pore

pressure buildup and cyclic strength reported by various authors is significantly reduced

or eliminated if pore pressures are compared for the same cyclic strain, rather than for

the same cyclic stress.

It appears from the results of this study that it will be possible to reasonably

predict pore pressure buildup and liquefaction potential of level saturated sandy sites

during earthquakes by means of a cyclic strain approach. This approach would be an alternative to the present state-of-the-art cyclic stress approach.

NOTE: The cyclic strain approach to liquefaction described in this paper was

originally proposed by R. Dobry from R.P.I, and was developed in References [3] and [9].

Further work was performed in the Summer of 197 9 as part of the earthquake engineering program of the Geotechnical Engineering Group of the National Bureau of Standards. This work

involved the participation of R. Dobry (RPI), F. Yokel (NBS) and R. Ladd (Woodward-Clyde

264
Consultants), as well as other staff members of the three institutions. The tests on

Monterey #0 sand presented herein were conducted at the soil dynamics laboratory of

Woodward-Clyde Consultants. The final report on this research effort [7] has been

published.

REFERENCES

[1] Seed, H. B., Soil Liquefaction and Cyclic Mobility Evaluation for Level Ground During

Earthquakes, Journal of the Geotechnical Engineering Division, ASCE, Vol. 105, GT2,

February 1979, pp. 201-255.

[2] Peck, R. B., Liquefaction Potential, Science Versus Practice, Journal of the

Geotechnical Engineering Division, ASCE, Vol. 105, GT3, March 1979, pp. 393-398.

[3] Dobry, R. , Ladd, R. S., Yokel, F. Y. , Chung, R. M. , and Powell, D. T. , Prediction of

Pore Pressure Buildup and Liquefaction of Sands During Earthquakes by the Cyclic

Strain Method, Building Science Series Report 138, National Bureau of Standards,

Washington, D.C., July 1982, 168 pp.

[4] Kovacs , W. D., Velocity Measurement of Free Fall SPT Hammer, Journal of the

Geotechnical Engineering Division, ASCE, Vol. 105, GT1 , January 1979, pp. 1-10.

[5] Park, T. and Silver, M. L. , Dynamic Soil Properties Required to Predict the Dynamic

Behavior of Elevated Transportation Structures, U.S. Department of Transportation

Report DOT-TST-75-44, May 1975.

[6] Yokel, F. Y., Dobry, R. , Powell, D. T. , and Ladd, R. S., Liquefaction of Sands During

Earthquakes - The Cyclic Strain Approach, Proc. , International Symposium on Soils

Under Cyclic & Transient Loading, Swansea, U.K., January 1980, A. Backema,

Rotterdam, Vol. 2, pp. 571-580.

[7] Dobry, R. and Ladd, R. , (1980) Discussion to Soil Liquefaction and Cyclic Mobility

Evaluation for Level Ground During Earthquakes, by H. B. Seed and Liquefaction

Potential: Science Versus Practice by R. B. Peck, Journal of the Geotechnical

Engineering Division, ASCE, Vol. 106, No. GT6 , pp. 720-724.

[8] Mindlin, R. D. and Deresiewicz, H. , Elastic Spheres in Contact Under Varying Oblique

Forces, Journal of the Applied Mechanics Division, ASME, Paper No. 53, November 1952.

[9] Dobry, R. and Swiger, W. F., Threshold Strain and Cyclic Behavior of Cohesionless

Soils, Proceedings, Third ASCE/EMDE Specialty Conference, Austin, Texas, September

1979, pp. 521-525.

[10] Dobry, R. , Powell, D. J., Yokel, F. Y. , and Ladd, R. S., Liquefaction Potential of

Saturated Sand - The Stiffness Method, Proceedings of the Seventh World Conference

on Earthquake Engineering, Istanbul, Turkey, September 8-13, 1980, Vol. 3, pp. 25-32.

265
1 10 100 1000 #5 000

NUMBER OF CYCLES TO FAILURE, n

Figure I Effect of fabric on cyclic strength in stress-controlled cyclic

tests (based on data from Park and Silver, 1975)

Figure 2 Effect of fabric and number of cycles on pore pressure buildup

in stress-controlled triaxial tests (based on data by Park and

Silver, 1975)

266
LEGEND

0 Crystal silica sand V - 14 psi Fresh samples, dry vibration

A Crystal silica sand V 14 psi Fresh samples, wet rodding

A Crystal silica sand 14 psi Staged testing, dry vibration

T Sand #1 - 20 psi Staged testing, moist tamping

V Sand 91 10 psi Staged testing, moist tamping

Figure 3 Measured pore pressure buildup in saturated sands after ten ioad

cycles (N = 10) of strain-controlled cyclic triaxial tests (Dobry

and Ladd, 1980)

267
0.06

0.05

0.04

^ 0.03

s

0.01

10'

oi = 2000 psf n = 10

Or Symbol

45 o

60

1 psf = 47.8 pascal

3 5 10

1

SHEAR STRAIN, 7, %

10

Figure 4 Residual pore pressure buildup after ten loa3ir>£> pyCics, straincontrolled cyclic triaxial tests (see aiso Yokel et al., 1980)

Figure 5 Effect of confining pressure, 533 psf <C 4000 psf

268
3000

1000

111

,1, w

*

/ —i i |

1 1

- Simple cubic array of quartz spheres

C^y*~ i

/

\2-v)[Uv)i , , 2/

/

/

for t = as

E = 11x10 6 psi

Kl /

l> = 0.31

7t = &76X10"5 (o)

2/

' %

-

i i l i _ —i 1Q- 3 3 5 10

=2 3 I 10"' 3 5 THRESHOLD SHEAR STRAIN, 7 t - %

ure 6 Calculated threshold shear strain as a function of confining pressure

for a simple cubic array of quartz spheres (Dobry and Swiger, 1979)

NT3 3 5 W~ 2 3 5 10' 1 3 5 1

SHEAR STRAIN, 7 %

Figure 7 Effect of number of strain cycles

269
Figure 8 Effect of relative density

270
THE PARTICIPATION FACTOR OF HORIZONTAL FORCE APPLIED TO PILE FOUNDATIONS

by

Yoshihiro Sugimura

Makoto Watabe

ABSTRACT

The participation factor of horizontal force applied to piles is discussed for the

purpose of rational seismic design of structures supported by pile foundations. Discussion topics are collected from various sources, viz., seismic damage investigation of

piles by the June 12, 1978 Off-Miyagi Prefecture earthquake, response analysis of an

example of a damaged pile foundation, earthquake observation data of fourteen story residential buildings supported on long piles and vibration tests for a model ground-pilebuilding system. A fundamental idea on the design of pile foundations is proposed as a

first step for a static method in which seismic effects are introduced.

271
INTRODUCTION

Present methods of evaluating earthquake loads on piles and the design concept itself

should be reconsidered because of the damage by the June 12, 1978 Off-Miyagi Prefecture

earthquake. Several examples of completely collapsed piles were found in Sendai City since

the earthquakes. This situation was the first full scale testing experience of buildings

supported on piles which extend through horizontal soil deposits in Japan. More research

on detailed on-site conditions through excavation will take place in the future as it will

produce valuable research data to improve the earthquake resistant design of pile foundations. Of course, a few examples of damaged piles were already recorded on the occasions

of the Niigata earthquake in 1964 [1,2] and the Tokachi-oki earthquake in 1968 [3,4], but

these are, more or less, due to differing causes such as liquefaction of sandy soil

deposits, or landslides on artificial fill ground.

Thus, research work has been initiated on topics such as the structural performance of

pile members, the effects of execution of work on characteristics of pile member, the

effects of horizontal loads on piles, the earthquake response of the ground-pile-structure

system, the earthquake-proof design method of piles, and the excavation and research on the

site mentioned above. The present paper is a preliminary report on these works and introduces the findings to date.

ACTUAL STATE OF DESIGN OF PILE FOUNDATION TO HORIZONTAL FORCES

The most customary design process to consider horizontal forces on pile foundations

for buildings is summarized in Figure 1. Horizontal forces at all floors are calculated

first under the condition of a fixed base. Secondly, by summing them and the moment

forces to the base, respectively, the so called base shear force, Q, and the overturning

moment force, M, are determined. After this procedure, the superstructure is removed and

effect of the foundation (footing and beam) is considered. Finally, the two types of

forces, divided by the numbers of piles, are obtained; i.e., the first is shear force

minus reaction force at the part of foundation from the base shear force, and the second

is the axial force which results from the foundation beam design which bears the overturning moment force. The effect of the axial force is generally considered in checking

272
for stresses in the pile members. This process may necessitate, at the least, resolving

the following questions

(1) Is it enough to consider only the inertia force of the superstructure as a

seismic design force on the pile foundation? Is it possible for the force of inertia to

be replaced by static force?

(2) Is it enough to consider just the component of horizontal force as a design force

at the pile cap?

(3) What percentage of base shear should be considered as a horizontal force

component at the pile cap? In other words, how large is the participation factor of the

horizontal force of the piles?

On the first question, it is necessary to consider the characteristics of ground and

superstructure in the frequency domain. The natural vibration modes of both can be

treated individually because both natural frequencies frequently disagree with each other.

From this fact, it is easy to estimate as follows: the vibrational characteristics of

the ground mainly control the behavior of the piles at and near the natural frequencies

of the ground. On the other hand, the vibrational characteristics of the superstructure

mainly control the behavior of the piles at and near the natural frequencies of the superstructure. The effects of the inertial force of the superstructure, therefore, predominant

at and near the natural frequencies of the superstructure.

Regarding the second question, it should be noted that the combined mode of horizontal

and rotational vibration is transmitted to the piles from the superstructure, and produces

a combined changed of shear stresses, bending moments, and axial stresses. In a strict

sense, therefore, it is necessary to take into account the effect of interaction between

factors such as axial stress and bending moment, axial stress and shear stress, and so on.

As for the third question, there are three types of resistance elements in foundations,

including basements if any exist, i.e., (1) subgrade reaction on front and back, (2) friction on sides and (3) friction on base. It is reasonable to take account of elements

(1) and (2), excluding (3), generally, due to lack of contact surface because of ground

settlement

.
PILE DAMAGE IN THE 1978 OFF-MIYAGI PREFECTURE EARTHQUAKE

Since the June 12, 1978 earthquake (Off-Miyagi Prefecture), several examples of

damaged pile foundations have been discovered in Sendai City. Table 1 shows the latest

confirmed pile collapses. The possibility of this type of failure was not apparent to

some researchers just after the earthquake. Gradually, suspicion increased as the effects

of differential settlement of the superstructures appeared, and it was finally confirmed

by excavation surveys of researchers such as the Sendai City authorities, Assistant Professor Kishida, Tokyo Institute of Technology, and others including one of the authors.

More research work on the actual conditions by excavation is planned from now on, because

many other examples are believed to lie concealed in this area.

This aspect of pile damage may be regarded, in a sense, as a historical event in

Japan. For this is probably the first time buildings supported by piles which extend

through horizontal soil deposits sustained damage from the vibrational effects of earthquake ground motion. The cases in Table 1, excluding case 5, all correspond to this

situation. The typical failure mode is regarded as bending, or bending-shear failure,

under axial load in the proximity of pile caps with and without final crushing. One case

is followed by complete collapse of superstructure, but others are followed by slight or

no damage to superstructures. These facts suggest the existence of a general tendency

that the stronger the damage to the piles, the slighter the damage to the superstructure.

The examples of damaged piles are shown in Figure 2 thru Figure 5.

The following factors should be considered at least a minimum list of the causes of

damage to the pile foundations.

(1) All the piles shown in Table 1 are concrete piles or prestressed concrete piles,

so that, it seems likely that decrements of strength to the pile members would be produced

in the process of execution of work completion, such as pile driving, or cutting off the

pile caps. As a matter of fact, many examples of small cracks, which suggest the after

effects of such work, were observed in other pile groups of a similar building which was

saved from final crushing. From this fact, particular notice in the designing process

should be taken of the difference between the piles as manufactured goods and the piles

after execution of work, especially prestressed concrete piles.

*

274
(2) The bearing capacity of a pile is designed considerably greater than the usual

design capacity as shown in Table 1. The safety factor of the piles might be smaller, as

the result of this fact. Also, a lack of consideration is recognized for pile member

characteristics when combined stresses due to vertical and horizontal forces are applied,

because the design calculation for horizontal force is not performed in other cases,

excluding case 3 and case 4.

(3) There is an indication that those buildings in Table 1 would be vibrated by a

larger earthquake ground motion than the design earthquake load of the standing code.

According to existing strong motion accelerograph records in Sendai City, the maximum

accelerations at the ground surface are within a range of 250 to 300 gals, varying of

course within each region. These values easily lead to the supposition that earthquake

loads larger than the design seismic coefficient K = 0.2 of the current code would act

on those buildings.

It is difficult and possibly meaningless to try to judge which factor is the

principal cause of pile collapse. Or rather, the hypothesis will examine the actual

condition sufficiently so that the safety factor designed at the first step would be

exhausted, because of overlapping of these disadvantageous factors to each other, and

then, collapse of piles would be produced finally.

ANALYTICAL EXAMPLE OF DAMAGED PILE FOUNDATION

An earthquake response analysis by the method shown in previous papers [5,6,7] was

performed for the building of case 3 in Table 1. The analytical models of the building,

the piles, and the ground are shown in Figure 6a to Figure 6c. One frame section in the

span direction (N-S) of the building, including the heavily damaged piles, was selected

as an analytical example and simulated by the model of 25 degrees of freedom, i.e., 11

masses in superstructure corresponding to each floor with single degree of freedom in

horizontal direction, 1 mass corresponding to foundation which has two degrees of freedom

in horizontal and rotational directions, and 12 masses in total in piles with single

degree of freedom in horizontal direction. It is necessary to divide the piles into two

groups because the extent of damage is clearly different between the heavily damaged

south side and the slightly damaged north side causing a building inclination of about

1/100. Six masses, therefore, are used per each group as shown in Figure 6b.

275
The input earthquake ground motion at the level of the pile tips was determined as

follows: the N-S component of the acceleration records observed in the basement of a

building located near the center of Sendai City was computed once to the tertiary bedrock

by the shear wave multiple reflection theory. By using that figure as the bedrock motion

of the site under consideration (GL-49.65 m in Figure 6c), the response at each level of

the soil deposits was computed by the lumped mass system. The maximum acceleration at

the ground surface and at the pile tip level using this approach to analysis are obtained

as approximately 430 and 370 gals, respectively. Then, the acceleration response at the

pile tip level is used again as the input earthquake ground motion for the analytical

models of the ground and the pile-structure system. Two cases, with and without rocking

of foundation, are analyzed.

All analytical model parameters are assumed to be linear elastic, in order to check

the point where stress failure is first produced. In a strict sense, a more detailed

investigation of the analytical models including the input earthquake ground motion will

be required in order to follow closely the failure phenomenon of the piles. Therefore,

discussions of responses such as acceleration and displacement at each mass are not

included here. Figures 7 and 8 show the time histories of the shearing forces and the

bending moments of the pile members at each level of the cases with and without rocking.

The number I corresponds to each number of the pile member shown in Figure 6b, so the

group from I = 1 to I = 7 shows the south-side piles, and the group from I = 8 to I = 14

shows the north-side piles. The arrow marks in these figures show the times when responses

first agree with the values of the failure shear stress and the failure bending moment of a

pile multiplied by the number of piles, i.e., four in the south-side pile group and five in

the north-side pile group.

It is recognized from these figures that the earliest point at which the responses

agree with the failure stresses is at the uppermost pile member of the south side (I = 7),

and earlier in the case of the bending moment than in the case of the shear stress. As

for the cases with and without rocking, the case with rocking is earlier than the case

without rocking. From this fact, it may be possible to suppose that the progressive

failure would be produced at the earliest failure point by the concentration of energy,

and finally, the crushing of the pile would appear around that place, while other points,

where the responses reach the failure stresses later, would be damaged only slightly arid

276
escape the final crushing. It should be noted, however, that: (1) the failure stresses

used here are estimated from the test results for the similar pile member not under the

axial load, so that the real failure stresses may be changing; and (2) in order to check

the real time of the failure and the behavior of the pile members after that time, the

analytical models should be more highly detailed and accurate.

EARTHQUAKE SPECTRAL RATIO BETWEEN PILE CAP AND PILE TIP

Figure 9 shows an example of the spectral ratio between the pile cap and the pile tip

of a building located in Tokyo, based on the seismic records from the past ten earthquakes.

Figure 9b shows the spectral ratio between the ground surface and the pile tip, based on

computed response at the ground surface using the seismic records at the pile tip as

input earthquake motions. Details concerning the earthquakes, the building, and the

ground condition were introduced in the previous paper [7]. The figures have been drawn

by changing the abscissa from frequency to period in the original figures in order to

clarify the differences of pile behavior between the natural periods of two ground and of

the building. Although these earthquakes belong to the category of destructive earthquakes, these figures explain the behavior of piles in the period domain very well. The

two spectral ratio peaks agree completely with the natural ground periods, i.e, the first

and the second shear vibration modes. Contrary to this, one of the valleys in the spectral ratio corresponds to the first natural period of the building. This fact suggests

strongly the possibility that the effect of the ground motion on pile behavior is much

larger than that of the superstructure.

VIBRATIONAL TESTS OF THE MODEL GROUND-PILE-STRUCTURE SYSTEM

Various tests using static horizontal force, free vibration, and the shaking table

were performed on structures (1 span - 1 story) with and without piles by the Committee

on the Horizontal Reaction of Piles. The Committee is headed by Professor Koizumi of

Tokyo Metropolitan University and includes one of the authors [8]. The main purpose of

the testing is to check the participation factors of the footing and of the piles under

dynamic conditions.

All model parameters, excluding the damping ratio, are adjusted by the similitude law

to the equivalent coefficients of the first vibration mode of a typical residential

277
building in Tokyo as shown in Table 2. The scale of length is selected as 1/30. Model

size and materials are as follows:

(1) ground (3.7 m x 1.8 m x 0.8 m, chemical materials with bentonite)

(2) pile (2 plates of 5 cm x 0.57 cm x 71.7 cm, steel)

(3) footing (box of 27.8 cm x 17.3 cm x 8.3 cm of 7.5 kg weight, steel)

(4) structure (2 columns of 5 cm x 0.2 cm x 17 cm of 12.4 kg weight, steel)

The layout of the test observation system is outlined in Figure 10a and 10b. Figure

11a and lib show a panoramic view of the model and its detail on the shaking table.

These show the model with 8.3 cm of earth from the surface, which corresponds to the

height of the footing. It is covered by a vinylon protection membrane.

Figure 12 shows an example of test results for the case of embedded footings. Figure

13 shows the distribution of the bending moment of the pile and the vibration modes of the

pile and the ground, and Figure 14 shows the earth pressure measurements at the footing.

The two clear peaks observed at the top of structure (B-ll) in Figure 12, correspond to

the natural frequencies of ground around 4 Hz and of the structure around 8 Hz, respectively. Noted, however, that the natural ground frequency would be changed to the small

frequency around 4 Hz from the first frequency near 5 Hz of the simple shear ground vibration mode from the effect of the wholly rotational vibration mode of the ground block,

and so on.

Location (G-10) is at ground surface removed from the model structure and not affected

by it. Response differences between footings (F-ll) and ground (G-10) reveals the sway

of the building so that one may divide the total base shear force into two categories as

follows

VG + VB (1)

in which,

an

E m.X. i=0 1 i (2)

an

V

E m X i=0 i G

(3)

G

278
nn

V,B

in which,

total base shear

Vq = base shear component corresponding to ground response

Vg = base shear component corresponding to relative response

m-£ = mass of each floor of building including foundation (i = 0)

Xg = acceleration response of free ground surface

nn = number of floors of building

Component Vg can be regarded as the inertia force on the superstructure due to the

excitation Xq. Component Vq is assumed to correspond approximately to the total result

of the interaction between ground and piles at each level. Results calculated using th

procedure and the ratio of each component to the total base shear force are shown in

Figure 15a and Figure 15b, respectively. These figures lead to the conclusion that the

ratio of inertia component of superstructure Vg becomes, approximately, 0.8 and 0.2 in

range of natural frequencies of superstructure and ground, respectively, and vice versa

for the ratio Vq. Figure 15c shows the ratio of the horizontal reaction at the footing

denoted by E to the base shear force component Vg, and vice versa for the shear force o

piles denoted by P, because the friction on the sides and bottom of the footing is negl

ted. The horizontal reaction at the footing is determined by the following algebraical

difference equation, based on the measurement by the earth pressure gages as shown in

Figure 14,

E = APi - AP2

in which,

APl = earth pressure increment at front of footing with positive sign

According to the vibration mode and the bending moment distribution observed on 3

accelerometers and 6 strain gages installed on the pile plane, as shown in Figure 13,

shapes similar to those obtained by static and free vibration tests of horizontal force

at the pile cap appear in the frequency domain around 8 Hz. However, the aspect in the

AP 2 earth pressure decrement at back of footing with negative sign

279
frequency domain around 4 Hz is quite different and strongly affected by ground motion

which is composed of the simple shear motion and the wholly rotational motion of the

model ground block. From this effect, the earth pressures measured at the front and back

of the footing include sizable components in the same phase with each other in the range

around 4 Hz, and a correction of removing these components is made in order to obtain

Figure 15c. It is necessary, therefore, to note that the result around 4 Hz does not

have the same degree of accuracy the one around 8 Hz. It is also necessary to note the

possibility that the earth pressure at the footing is acting as a kind of input force

rather than as a reaction in the range of about 6 to 7 Hz in Figure 15b or Figure 15c.

TENTATIVE DESIGN OF PILE FOUNDATION TO HORIZONTAL FORCE

From the above discussion, a simplified draft procedure for the design of pile

foundation to horizontal force is proposed as shown in Figure 16. The principle of this

procedure consists of dividing the external design force into two categories, i.e., (1)

the usual concentrated horizontal force at the pile cap which is determined by the inertia

force of the superstructure minus the reaction at the foundation, (2) the distributed

horizontal force over the total length of the pile which is determined by distributing

Vq, for instance, into a triangular shape. The inertia force of the superstructure can be

determined by assuming some available distribution shape of the horizontal force on each

floor, and, of course, the sway component at the foundation if necessary. As for the case

of total inertia force, it is considered to already include the effects of the underground

part, and dividing it into two forces using the relation shown in Figure 15b is recommended.

The design participation factor of the foundation reaction should be taken at an appropriate value from the relation shown in Figure 15c, and a value of not more than 0.3 is recommended here. Regarding the distribution of Vq, the most fitting shape, triangular or other,

should be used to represent the effects of the natural ground vibration modes. The effects

of grouping piles also should be taken into account as required. Finally, it is necessary

to take the sum of the pile stresses from both branches to design the pile section and

footing. At least two cases must be checked, i.e., at the first natural frequency of

the superstructure and the ground; then design for the most disadvantageous case.

280
CONCLUSION

A reconsideration of the current static design method which conceptualizes the

horizontal force being concentrated at the pile caps is strongly suggested by the observed

aspects of earthquake damage to piles, and results from the model vibration tests discussed

in this paper. The more desirable method is the dynamic response analysis of the total

system of ground, pile and superstructure to earthquake ground motion. However it would be

very convenient in the earthquake-resistant design of buildings supported by piles to have

an appropriate static design method for piles which pays due regard to the seismic behavior

of the ground. A tentative draft of such a method is proposed in this paper which includes

the participation factor of horizontal force of piles. More detailed research on the design

method for piles will be continued, however, based on the topics mentioned in the first

sections of this paper because the problem of how to deal with the distribution of external

force over the pile length, the effect of earth pressure in the input force state, the

effect of pile grouping, and so on, still remains.

ACKNOWLEDGEMENT

The authors wish to express their gratitude to Assistant Professor H. Kishida, Tokyo

Institute of Technology, the Sendai City authorities, and many other researchers for providing earthquake damage data on the piles, and also to Professor M. Izumi, University of

Tohoku, for the use of some subroutine subprograms coded by him and his colleagues. The

authors also wish to thank Professor Koizumi and all the members of the Committee on the

Horizontal Reaction of Piles and especially the members of its working group for frequent,

stimulating and helpful discussions.

281
REFERENCES

[1] Kishida, H. (1966), "Damage to Reinforced Concrete Building in Niigata City with

Special Reference to Foundation Engineering," Soils and Foundations , Vol. 6, No. 1, pp. 71-88.

[2] Fukuoka, M. (1966), "Damage to Civil Engineering Structures," Soils and Foundations , Vol. 6, No. 2, pp. 45-52.

[3] Tamura, K. , Fujiwara, T. and Morishige, R. (1973), "Damage of Rigid Frame Bridges in

the Off-Tokachi Earthquake 1968," Japan Earthquake Engineering Symposium 1973, in

Commemoration of of the 50th Anniversary of The Kwanto Earthquake, Tokyo, Japan,

pp. 147-154 (in Japanese).

[4] Building Research Institute (1974), "Report of Aseismic Diagnosis on the Reinforced

Concrete Building of Aomori Prefectural Business High School," Report No. 206 (in

Japanese)

.

[5] Sugimura, Y. (1973), "Estimation of Dynamic Behavior of Soil Layer-Pile Foundation

Interaction System During an Earthquake," Japan Earthquake Engineering Symposium 1973,

in Commemoration of the 50th Anniversary of The Kwanto Earthquake, Tokyo, Japan,

pp. 133-140.

[6] Sugimura, Y. (1975), "Earthquake Observation and Dynamic Analysis of Pile-Supported

Buildings," 4th Japan Earthquake Engineering Symposium, pp. 463-470.

[7] Sugimura, Y. (1976), "Earthquake Observation and Dynamic Analysis of a Building

Supported on Long Piles," 6th WCEE, Vol. 2, pp. 1570-1575.

[8] The Committee on the Horizontal Reaction of Piles (1979), "Report of Study on Seismic

Passive Earth Pressure to Foundation and Beams of a Tall Residential Building," (in

Japanese)

.

282
Table 1 Damaged Piles by 1978 Off-Miyagi Prefecture Earthquake

No. Type of

Bui 1 di ng Pile L (m) D (cm) Design Bearing

Capacity (t/pile)

Fai 1 ure TypeDamage of Structure

Geographical

Condition

Main Soil Type

1 Box-RC 3FL RC 5 25 20 BB CO Plain Peat, Clay

2 Box-RC 4FL PC 5 35 70 CR NY Plain Clay, Sand

3 SRC 11 FL AC 12 60 150 CR SY Plain Silt, Sand

4 SRC 14FL AC 24 60,50 160 CR SY Plain Peat, Sand

5 RC 4FL PC 10 30 30 BS SY Slope Loam

Notation: RC=rei nforced concrete pile PC=prestressed concrete pile AC=autocl aved

prestressed concrete pile L=average length of pile D=diameter of pile

BB=bending cracks without crush BS=bending-shear cracks without crush

CR=crush with bendi ng-shear cracks C0=col lapse NY=no damage with

differential settlement SY=slight damage with differential settlement

Table 2 Properties of Prototype Structure and Model Structure

under Rules of Similitude

Unit Prototype Model

Weight of Structure kg 446700 12.4

Peri od of Structure sec 0.63 0.12

Weight of Footing kg 2.7x10 s 7.5

Size of Footing cm 835x520 27.8x17.3

Depth of Footing cm 250 8.3

Diameter of Pile cm 150 5x0.57*'-

Geometrical Moment

of Inertia of Pile cm

1* 2.5xl0 7 0.0769

Depth of Ground cm 2400 80

Period of Ground sec 1.10 0.21

Shear Wave Velocity m/sec 91 16.75

Model of pile is replaced by rectangular shape section.

283
0 = I Ff Qp = (Q-Qf) / n

M= I F,H,

F| em|X| (Sustained by

Foundation

Beam)

M

N : Considered at Final Check of Pile Stress

Numbers of Pile = n

Fig. 1 Actual State of Design of Pile Foundation to Horizontal Force

Fig. 2a) Damage of Pile (Case 1) Fig. 2b) Damage of Pile (Case 1)

284
Fig. 4a) Damage of Pile (Case 3)

285
Fig. 5a) Damage of Pile (Case 5)

Fig. 5b) Damage of Pile (Case 5)

286
Fig. 6a) Analytical Model of Ground and PileStructure System (Case 3)

00

4:6 5_

_S_65_

6^65

7 65

Fig. 6b) Details of Model of Piles and

Ground (Case 3)

m

1165

South North

11S5

990

8551 .84lA® A©

_LCL

®

6 0 9

5 Q -e20

O

<2> ®

m

0 0

CD ®

16

287
0 0 < 16

1 75 C 15

3 65 C 4 .65 C H3

5 65 C ^12

6 65 C

i i 7 65 ( ^10

9 65 < 3 9 n 65 ( 8

12 65 C 7

16.65 6

19.65 9,

26.65 9

36.15 9

T hi r k n p ^ ^ (m)

i 1 T vnp Unit Wei ah

t

(t/D 5

)

N Value ^hpsr find 1 1 1 tic

( t/m2

)

^.' hps l i r IJ a up Vol nri t\/ C O i Ha VC « » 1 Ut 1 Lj (m/sec)

3.65 CC 1 . 50 4 3637 .

0

1.0 SS 1.85 9 6959.5 192.0

1.0 SG 1 .97 31 15718.4 305.2

1 .0 52 28312.9 375.3

1.0 " 83 41157.0 458.5

4.0 II Ii 44 24770.0 351.3

1.0 Ii ll 99 47390.3 485.5

4.0 It 49 26998.

4

366. 5

3.0 »' II 77 38760.0 439.5

7.0 II II 79 39560.0 444.0

19.0 II 96 46240.0 480.0

2.0 II H 19 12650.0 251.0

2.0 II 94 45470.0 476.0

RR 2.20 200 83177.4 608.7

Fig. 6c) Ground Condition at the Site of Case 3

288
Fig. 7 Time Histories of Shear Stress and Bending Moment of Piles

(Case with Rocking)

1*10' 1 I ,»t

~i ( SEC I

' 6HEAR S1RE66 OF PILE 1- 7

r°-m i ( sec ) SHEAR G THESE OF PILE J- 8

I , --jVW"-\/i ! ( SEC 1 SHEAR STRESS OF PILE !- 5

'jvyj i ( SEC 1 SHEAR 6TRES6 OF PILE J- 1

1 Jt —

i

6 - SHEAR S1RL6S OF PICE 1-lM

[»10« I —i'^W^sJi « ( SEC I 6HEAR SIRENS OF PI E J - 1

3

TT^XP i ( SEC I SHEAR STRESS OF PILE 1-12

3 61/* 'iVVj i ( SEC 1 6HEAR STRESS OF PILE J-ll

1

i ( SEC I SHEAR 6TRES6 OF PILE I- 3

~i I SEC ) SHEAR STRESS OF PILE 1-10

10. , fr/v^^yi—iisEci

J— r^T^rs^p i ( SEC 1 10, SHEAR STHESS OF PILE I- 2

1—--'(V'v/'J \Ji J ( SEC ) l0- ' SHEAR STRESS OF PILE I- S

6HEAR STRESUT\Ji 2 ( 6EC ) iU l)F PILE 1- 1

' 6HEAR SIRE66 OF PILE I- 8

10.

T 1 1 6EC 1 HOMENT OF PILE J- 7

~i ( SEC 1 MOMENT OF PILE 1- 8

MOMENT uF PILE I

1 1 ( SEC )

3* 1 °" 1 ' r

|—i

"

fvW\/i

i(l

3 ' MOMENT OF PILE I- II

I I SEC ) MOMENT OF PILE I- 3

i "v lVVi \n I I SEC 1 MOMENT OF PILB* I- 2

i T^V \ ~i ( 6EC 1 MOMENT OF PILE J- I

1

i I SEC )

—i—-fvAxw;—

i

MOMENT nF PILE j-i;

-—

IF PILE J - 1 I

t I SEC )

-fv^/1'VY')

1 | SEC ! MOMENT OF PILE 1-10

i ""ft/Vi \/( j ( SEC ) MOMENT OF P1LF 1- 8 —i—•y'VV-^ ? MOMENT OF PILE I- 8

^89
F1g. 8 Time Histories of Shear Stress and Bending Moment of Piles

(Case without Rocking)

—i -~r/v>' SHEAR STRESS

a"V

rp 1 (stci

: or pile j- 7 —i

SHEAR STRESS*\fi l (SEC)

fr pre i- e

AR STRESS OT PILE ]

6EC 1

5

3-

10.,

10.

—I ( EEC ) SHEAR STRESS OF PILE J- 1

1 —r—t~-"{\j<( 1 1 sec ) SHEAR STRESS OF PILE ]- 3 —;—-T T\fl 1 ( SEC 1 SHEAR STRESS Or PILE 1- Z

tffl 1 t SEC 1 SHEAR STRESS Of PILE 1- t

—r-^/V)

SHEAR STRESS*d~\—i ( sec i pF PILE I'll —i

-T-Myi ? ( sec ) 6HEAR 6TREB6 tF PILE 1-13

^»ArHA u

R 61RE6B OT PILE 1

SEC ) -12

^^r\f(—1 1 sec t SHEAR STRESS Or PILE 1-11

1 —\-'-~r'-'f\Ji{ 1 1 sec i SHEAR 6TRE6S OF PILE 1-10

i J" f\/i 1 I SEC ) 6HEAR STRESS or PILE I- 8 —I—

?

T\f\ 1 ( SEC ) SHEAR STRESS tlf PILE I- 8

i * i u

1

j

> ¥l-S^r\f\ 1 I SEC I l0- HOHEN) Or PILE 1- 7

L^tO'l

j—i -t^Jr\f\ MHMrwi nc p 1 1 c iI 1 SEC) MOMENT Or PILE 1- 8

MOMENT IF PILE I- 11

I 1 0 )

1

I I 6EC ) B - MOMENT OF PILE I- 3

,0- MOMENT or PILE I I

D

1 OF PILE 1-1

MOMENT Or PILE 1-12 MIC )

i I SEC I l0 ' MHMENT Or PILE 1-11

1 ( SEC ) 8" MTMENT Or PILE )-IO

— 7 f\fi J16EC)

10- ' MOMENT OF PILE I- 2

t «10« 1

1

i

' —y"^^/\> 1 ( SEC ) '• MOMENT Or PILE 1- 1

MOMENT Or PILE I- B

I v- I

j

1 'j^V-A^ J ( 6EC 1 MOMENT Or PILE 1- 8

290
! - M ! p

f F P !

! ! Ij f

• t !

9a) Spectral Ratio Between Pile Cap and Pile Tip

Based on Seismic Records

Fig. 9b) Spectral Ratio Between Ground Surface and

Pile Tip Based on Analytical Results

291
©-B-11

F-13

Building.

B

E-12

>0 ie-o r120 4— S-3 0 120 |- S-2<

160

| S-lfl 170 JL_

ML

s-5 <f- -e- - G-21 e- - p-i

G-22 P-2

P-3

200

20-0

250

T-l

500 -e>B-2i

Boilding.A

F F-21 " 2 I

) . -e» j

•e- - G-31

G-32

T-l

RE-21 8-3 y§7

200

45-0

W 1-70 o

HS.C 1 !•»*-

ae' aC

Unit cm

e- Accereiometer

c Strain Gag?

D Earth Pressre Gag*

Fig. 10a) Layout of Accelerometers , Strain Gages and

Earth Pressure Gages

500 AA

to o BA

i-CA.

SOO

95-0'

L,

l-7oa-

-H5.0

-AB' AC Unit : cm

185 0 1

-900-

10

•50 0- 60 0- -50 0-

80 0-

Building A

80 -0—

Building,

B

&H0

G11

15-

20-0 ^G12 41

G13

2j>.0

25-0

S3

717

&^31 1rfl rG*5 A.

G-32 0-12 &-22

G-13

1 1

80-0

Fig. 10b) Outline of Location of Model Structure

292
Fig. 11a) Panoramic View of Mode

Fig. lib) Models of Structures

293
Fig. 13 Cencling Moment of Pile, and Vibrational

Mode of Pile and Ground

294
Fig. 15 Base Shear Force Component and Participation Factor

Z95
Fig. 16 Draft Design of Piles

( Start )

^ Assumption of acceleration response X~ at ground surface

i

Calculation of base shear force

component with fixed base

V D = Zm.X®

Calculation of Qj.

Qp 4vB (l-a)

design particioation

factor of footing

numbers of pile —*fi

Calculation of horizontal

resi stance

Calculation of base shear force

component corresponding to

V r

= Em.jL

b 1 b X

Sharing with depth

Q, = (L-z.)V

2 1 rt> nL s

total length of pile

depth of i th level

Calculation of horizontal

resi stance

Summing up stresses of pile

Design of pile section

I Design of footing

Qnd~)

296
OPERATIONAL MODELS FOR FORECASTING STORM SURGES IN NOAA

by

Celso S. Barrientos

The National Weather Service of the National Oceanic and Atmospheric Administration

(NOAA) is responsible for forecasting in the air and water environment. Operational models

are those that are available for use at regular and scheduled basis. For example, the

operational 7-level primitive equation model of the atmosphere is run twice a day everyday.

There are many models in NOAA and among these are operational models. This paper is concerned only with the operational models for forecasting storm surges in NOAA. These are

other storm surge models both inside and outside NOAA but they are not operational. Storm

surge is the abnormal rise (or fall-negative surge) of water level due to meteorological

forcings. The regular fluctuations of sea level (tides) are not included in storm surge.

There are three areas in the United States where storm-generated water level fluctuations

(storm surges) can cause serious problems. There are: (1) in the Great Lakes, particularly

Lake Erie; (2) in the East coast of the U.S. —winter storms in the northeast portion, and

tropical cyclones or hurricanes throughout the coast; and (3) in the Gulf of Mexico coast

due to tropical cyclones or hurricanes.

Storm surge can cause considerable damage so that timely and accurate forecasting is

essential. In Lake Erie, flooding and shoreline erosion occur when positive surges are

superimposed on high lake levels. During times of low lake levels, negative surges can be

hazardous to navigation and disturb hydroelectric power generation. Similarly, low water

level is also a problem in Baltimore Harbor, Maryland, because it causes the ships to scrape

the bottom. Storm surges in the Gulf and East coasts of the U.S. cause considerable damage

every year. Storm surge episodes along the coasts always leave phenomenal amounts of property damage and a frequent loss of lives. In recent years, timely warnings have helped

decrease the losses and have saved lives.

Storm surge is generated mainly by two atmospheric forces —wind and pressure. Wind

acts on the water with a drag force (stress) and low pressure causes the water to bulge

(increase elevation). The low pressure force is also termed the inverted barometer effect.

These two forces act in a dynamically coupled way and must be considered together. Storm

surge forecasting therefore, must begin with forecasting atmospheric forces. Different for297
mulations of atmospheric forcings are used with the different storm surge models. Success

or failure of a storm surge model depends very strongly on formulation of the meteorological

forces

.

In the Great Lakes, we will focus our discussion on Lake Erie. There are other

locations in the Great Lakes where storm surges do occur; however, the frequency and magnitude are comparatively small. On the average, major surge (3 ft or higher) occurs in Lake

Erie about five times a year.

Lake Erie storm surges are generally associated with the passage of extratropical

storms through the Great Lakes area. Winter storms that approach the Great Lakes from the

central part of the United States can cause strong winds over Lake Erie. When these strong

winds blow along the main axis of the lake, storm surges are generated. Strong southwest

winds blowing over the lake cause a tilted lake surface, i.e., the water level is elevated

in the eastern portion of the lake (Buffalo, N.Y. end) and lowered at the western portion

(Toledo, Ohio end). Occasionally a storm will pass south of Lake Erie causing northeast

winds over the lake. When this happens, the slope of the water surface is reversed

with elevated water levels at the Toledo end of the lake.

Storm surges on Lake Erie are caused primarily by wind stress on the lake surface. The

effect of atmospheric pressure, which causes higher water levels in areas of low pressure

(bulging effect) is less important. The reason is the lake is relatively small compared to

the size of the storm. Storm surges in Lake Erie are pronounced because of the shallow

depth and the geographic orientation.

The National Weather Service of NOAA is responsible for predicting storm surges in

Lake Erie and warning all those who may be affected to take necessary action. The first

operational model was implemented in 1969 and it is based on equations derived by statistical techniques. Regression equations were derived from historical sea level pressure

data around the Great Lakes. The predictand used is the actual measured lake levels or

storm surges. The statistical model has worked with a high degree of success for ten years.

Objective methods to forecast winds over the Great Lakes became available in 1972.

With the wind forecasts, a dynamic model to forecast storm surges in Lake Erie was developed.

It uses impulse response functions to calculate the storm surge height. The surge height

at a given time is calculated as the weighted sum of forcing terms during some period before

the specified time.

298
Using two years of data (1977 and 1978), the statistical and dynamical storm surge

models for Lake Erie were compared and verified. Results of the study showed that the

dynamical model is better than the statistical model. Therefore, the operational statistical model was replaced by the dynamic model in 1979.

In the East and Gulf coasts of the United States, storm surge generation is further

enlarged by additional factors. These are water transport by waves and swell, the Earth's

rotation, rainfall, and coastline configuration and bathymetric conditions. Theory and

laboratory experiments indicate that waves breaking near the shore contribute to the storm

surge. Wave setup is also affected by wave refraction—that is, wave setup is higher in

areas of wave convergence and smaller in areas of wave divergence. The Coriolis effect

(earth's rotation) will increase the surge height on the coast when current is parallel to

the coast and the coast is to the right of the direction of the current; the reversed effect

is true for current moving in the opposite direction. Hurricanes and extra-tropical storms

may produce a great deal of rain over extensive areas, resulting in floods that can increase

the surge height near the mouths of rivers and tidal estuaries. The bottom topography near

the shore has an extremely important effect on the height of the storm surge. Gently

sloping topography on the continental shelf, as in the Mississippi river delta, supports the

generation of higher storm surges than does a steep continental shelf. The configuration of

the shore also affects the storm surge. For example, storm surge height increases in a bay

with converging shorelines but decreases in a wide bay with only a narrow connection to

the sea.

The path of extratropical cyclones is along the eastern coastline of the United States.

This path is favorable to surge generation in coastal states from Georgia to Maine. The

National Weather Service Forecast Offices are responsible for issuing storm surge warnings.

Operational computerized statistical forecasts are used as guidance at these forecast offices. This automated forecast technique, which became operational in 1971, is a statistical

model based on actual November through April storm surge data, from 1956 through 1969, using

storms that produced surges of 2 ft (0.6 m) or more. Separate forecast equations were

determined by the statistical screening procedure for each of the forecast locations so that

local effects were considered.

In actual operational use, sea level pressure forecasts at the appropriate grid points

are inserted in the storm surge equations. Pressure forecasts are made twice daily from

299
numerical weather models in the Weather Service's National Meteorological Center. This

model will continuously be improved by including recent extra-tropical storm surge data

beyond the original 14-year development information. More forecast locations will be added

to the system, which is strongly dependent on the atmospheric models forecasts. However,

during the nine years of use it has performed satisfactorily.

Hurricane storm surge produced considerable damage in the U.S. coastal areas. NOAA has

two operational models to forecast hurricane storm surges —a model for the outer coast only

and a model for bay and estuary for overland flooding. These models are dynamic. The storm

surge dynamic equations are solved numerically. The storm models include an atmospheric

model. Simple variables to define storms are forecast and used as input to the models.

These variables include the central pressure of the storm, the size of the storm, the direction, and speed of motion. The outer coast model uses a curvilinear coordinate system

following the coastline and a variable grid toward the sea, which is a function of depth.

The bay model employs a polar coordinate system, with a smaller grid for bays which expands

towards the ocean. The bay model is dynamically coupled with the ocean model, thus no input

boundary is required. It is operational in five areas and will be implemented in 20 important bay and estuary systems in the Gulf and East coasts. The hurricane storm surge model

mechanisms are run only when a storm is threatening the U.S. coast. These models threatening the U.S. coast. These models provide valuable information to forecasters in the preparation of the storm warnings, and are also used in evacuation planning.

300
STORM SURGE PREVENTION AROUND RIVER MOUTHS IN JAPAN

by

H. Hashimoto

ABSTRACT

To prevent storm surge damage, defense work has been carried out since the Isewan

Typhoon caused heavy damage in 1959. Around river mouths, dikes have been constructed for

the protection of densely populated areas, the dimension of dikes were determined with

considerations of storm surges and high waves. Prediction models for surges and hydraulic

models for waves provided necessary information. However, the effect of dikes was weakened

by land subsidence caused by ground water pumping. The stability of structures against

earthquakes is another problem. Efforts continue to develop new methods for the improvement

of structures.

301
INTRODUCTION

Japan is located on the western side of the Pacific Ocean where typhoons strike about

seven times a year between July and September. In a bay, plains are low flat deltas formed

by the deposit of sand and silts transported by rivers. Population and structures are

concentrated in plains such as Tokyo in the Kanto plain.

Once storm surges causes flooding in a delta area, damage to life and property can

become enormous. As one countermeasure , dikes have been constructed along the coasts and

river mouths. This report describes the design method for those structures and newly

arising problems.

BACKGROUND

When a typhoon hits a bay with a mouth opening to the south, storm surges occur.

Figure 1 shows frequently hit bays, and tracks of typhoons which caused large surges. In

the delta area of these bays, land subsides from ground water pumping. Figure 1 also shows

areas where the ground is lower than the mean sea level. These areas are protected by

coastal and river dikes from flooding. Once storm surge breaks a dike, drainage of the

accumulated sea water is very difficult.

In 1959, Isewan Typhoon hit Ise Bay and caused heavy damage in Nagoya. Five thousand

lives were lost and property loss was estimated at five hundred billion yen. At that time,

prevention measures against storm surges were reconsidered and a new technique was developed.

For storm surge defense works, dikes were constructed along the coast and river mouth in

Tokyo Bay, Ise Bay and Osaka Bay.

DESIGN METHOD

Dikes or breakwaters prevent flooding from surges and high waves. Because a flood from

the upper river must flow easily, it is not desirable to construct structures which totally

obstruct the stream. Therefore, dikes are used with gates and pumps which drain the low

land

.

In harbor areas, because high dikes hinder the use of quays, the flooding is prevented

by breakwaters. At Nagoya harbor, 8.2 km long breakwaters were constructed to protect the

302
harbor area from storm surges and high waves. The breakwater has two openings, 350 m and

50 m which decrease the surge from 3.5 m to 3.0 m.

The dike was designed using the following procedure [1]:

(1) establish a design typhoon

(2) estimate the height of storm surges and wind waves at the mouth of the river

(3) analyze the deformation of the surges and waves in the river

(4) estimate the wave run-up height on the dike

(5) decide on the type and the dimensions of the dike

Storm surge and wave height are estimated from numerical computation. Deformation and

run-up of waves are analysed using hydraulic models. The dike is designed to be higher than

the wave run-up height. In this case, the water level is determined by the surge height

and the mean high water of spring tides. The dike is covered using concrete or asphalt for

stability from attack by waves and tides.

The storm surge defense work began in 1957. In 1959 the Isewan Typhoon caused heavy

damage. Then the design typhoon was determined to have the magnitude of the Isewan Typhoon.

The lowest pressure at the center of the typhoon and the maximum wind speed show that the

Isewan Typhoon is one of the largest typhoons that we have experienced.

The analysis of construction cost and benefits based on statistical data can be used to

determine the dimension of the dikes. However, this method was not applied because of a

lack of adequate data. At Nagoya, the anomaly of storm surges was analysed from a total of

20 years data [2]. The return period of the anomaly caused by the Isewan Typhoon ranges

from 40 years to 70 years.

STORM SURGE COMPUTATION

In estimating surge height, the design typhoon moves on a track which produces the

highest surge at the place where it is to be examined. The track is selected to be similar

to the ones of the past. When the surges were estimated at the mouth of the Ibi River,

Nagara River and Kiso river at the bottom of Ise Bay, five tracks were selected [3]. Figure

2 shows three of the five main tracks. Track A is a simplification of the Isewan Typhoon,

and B is moved 20 km to the west and parallel to A.

Storm surges are computed by two-dimensional long-water wave equations which consist of

momentum and mass conservation equations [4], The pressure distribution in a typhoon is

303
derived from observed data. The wind on the sea surface is assumed to be the vector sum of

the velocity derived from the atmospheric pressure gradient, and the velocity proportional

to the typhoon speed. The design typhoon produces the wind distribution as shown in Figure

3. The tangential force on the sea surface caused by the wind drives the water and generates the surges. The computation is carried out on a grid of several kilometers with about

one-minute intervals. In the case of Ise Bay, a wide grid of 3.2 km and a narrow grid of

0.8 km were used. The deformation of surges in a river are also computed by one-dimensional

long-water wave equations.

As an example, the time history of sea levels at Nagoya, Kuwana, and the mouth of the

Ibi River is shown in Figure 4. (The location is also shown in Figure 5.) A comparison

between observed and predicted levels are also made, as is the highest level at each station

produced by the different tracks shown in Figure 3. At the mouth of the Ibi River, the sea

level reaches 4.8 m above mean sea level.

WAVE RUN-UP ESTIMATION

Because storm surges accompany high waves, we need to estimate the wave run-up on dikes.

The wind waves generated by the typhoon are computed by Wilson's moving fetch method [5]

based on the SMB method. In shallow water, energy dissipation from bottom friction is also

considered by Bretchneider ' s equation [6].

At the mouth of the Ibi River, five fetch lines were selected. Waves from each

direction were calculated at 5 km intervals [7]« Figure 6 shows the time history of significant wave heights and periods from each direction in front of the river delta. The maximum wave is 5.5 m high and of a 7.9 sec period. At the river mouth, it decreases from

bottom friction to 3.0 m and 7.5 sec.

Waves which reach the river mouth deform by refraction, reflection, dissipation, and

run-up on the dike. The deformation is investigated using hydraulic scale models to which

the Froude similarity law is applied.

The model scale is usually from 1/30 to 1/60. The bottom friction coefficient in the

prototype is estimated to be f = 0.01, but in the model, the coefficient is about 0.04.

From the scale law, they should be the same. Therefore, some corrections are required to

the experimental data. A distorted model is also applied; and, though the correction is

small, sometimes wave generation by wind on the river is considered.

304
At the mouth of the river, waves deform according to bottom profiles. Because dredging

and reclamation is active in the delta, we need to assess the influence of the modifications

on the bottom profiles. A gradual change of the bottom profile from a deep center of the

river to a shallow bank, is preferable because the wave refracts to the banks and does not

enter the upper reaches.

In the model experiment, wave height in a river and run-up height on a dike are

measured to estimate the required dike height. As an example, Figure 6 shows the change of

run-up height in the Ibi River [8]. The experiment was carried out in a horizontal scale

of 1/65 and a vertical scale of 1/32.5 to analyze several methods to dissipate wave energy.

In the figure, a comparison is made between the present dike and the dike with a berm.

Waves break on the berm and both wave run-up and transmitted waves to the upper river

decrease. For the prevention of storm surge disasters, the dikes have been designed as

outlined above and constructed in three major bays.

ADDITIONAL PROBLEMS

As measures for the prevention from flooding coastal and river dikes were taken,

however, a few new problems arose. They are:

- land subsidance caused by ground water pumping

- instability of structures on weak soil against earthquakes.

In Nagoya, the land subsides 3 to 10 cm a year and the area under the mean sea level

gradually increased from 180 km^ in 1959 to 250 km^ in 1977 as shown in Figure 5.

The dike also sinks from land subsidance and from a consolidation of the subsoil. The

latter becomes inactive as time advances.

The pumped water is used for industry and agriculture. At the Ibi River, tide

influences areas over 30 km from the mouth and salt wedges interfere with the use of river

water. Deep wells supply water to rice field in the delta area. There are five layers at

the mouth of the Ibi River and they are: upper gravel, upper clay, lower gravel, lower

clay, and the tertiary from the surface to the bottom. The pressure depression in the

lower gravel caused by the pumping causes a consolidation of the upper clay by the excess

pore pressure.

At the Ibi River the dike was reconstructed because of the land subsidance. The river

bed also sinks and waves become higher than before. Several measures were investigated

305
experimentally to reduce waves [9]. Figure 8 shows the effect of berm width on wave run-up.

As the berm becomes wider, the run-up decreases because of the wave breaking on the berm.

However, a berm wider than 15 m is not effective. Figure 9 shows the actual design of

dikes on the Ibi River.

The stability of vertical walls which were constructed in densely populated areas is

the other problem. Once the wall is broken by an earthquake, inundation on low land causes

heavy damage. Efforts are continuing to make the structure stable. In Tokyo there is a

plan to raise the land higher than mean sea level, and to make a wide dike as shown in

Figure 10. Dikes covered with green grass or natural stones are preferred to high concrete

walls

.

SUMMARY

After heavy damage caused by Isewan Typhoon, dikes have been constructed around river

mouths in delta areas. The dimensions of the dikes are designed to withstand storm surges

and waves generated by a design typhoon which has the same magnitude of Isewan Typhoon. To

estimate the height of surges and wave run-up, computational and hydraulic models have been

used. In Tokyo Bay, Ise Bay, and Osaka Bay the defense works were completed. However, the

effect of the dikes was weakened by land subsidance caused by ground water pumping. The

stability of structures against earthquake is another problem. Efforts are continuing to

develop new methods to improve these structures.

306
REFERENCES

[1] Design Manual for Coastal Protection, Association of Coastal Engineers in Japan, 1972.

[2] Report on the Analysis of Storm Surges in the Three Rivers of Kiso, Kisokaryu

Construction Office, Ministry of Construction, 1971.

[3] Ibid, [2].

[4] Miyazaki, M. , S. Unoki and T. Ueno, "Computation of Storm Surges Caused by the Isewan

Typhoon," 7th Coastal Eng. Conf. in Japan 1960.

[5] Wilson, B. W. , "Graphic Approach to the Forecasting of Waves in Moving Petches," BEB. Tech. Memo., No. 73, 1955.

[6] Bretchneider , C. L. and R. 0. Reid, "Modification of Wave Height Due to Bottom

Friction, Percolation, and Refraction," BEB. Tech. Memo., No. 45, 1954.

[7] Ibid [2].

[8] Model Experiments of the Storm Surge Prevention Structures in the Three Rivers of Kiso,

Tech. Memo., PWRI, 1973.

[9] Hydraulic Model Experiments of Dikes in the Three Rivers of Kiso, Kisokaryu

Construction Office, Ministry of Construction, 1979.

307
(1951) No. 16 (1942)

Fig. 1 Tracks of Violent Typhoons

308
4

Tracks of Typhoons in Storm Surge Computation

309
Direction of Typhoon

Fig. 3 Distribution of Speed and Direction

in the Design Typhoon

310
cm 500

0 I i | | | | | | | | | I I I

| 10 12 14 16 18 20 22 24hr

Time

Time

Fig. 4 Storm Surge caused by Design Typhoon

311
Fig. 5 Map of Ise Bay

312
Sec

8.0

1 6.0

Wave Period

>

:* 4.0

• o

• o »

• o

o * o

O

O

a

• o

o " o 2.0 -

Fetch Line

o n

• m

» W

J I I L J I L 15 16 18 19

Time

2 I 22hr

m6.0

-p

•f 4.0

•H

<D X

Wave Height <L ^ • 9

3 O - • • O O 9

9

• O

2.0

Fetch Line

o n

• m

j | | | I I I I I I I I I I I 15 16 17 18 19

Time

20 21 22 hr

Fig. 6 Change of Vave Height and Period

313
• Present Dike

o Dike with a berm

• | o • 8-5— — —1 • Design Height

-

"—5- o o

o *

o o o

Left Bank

(Nagara River)

J I I I L 0 2 4 6 Distance from the Mouth

-p-H

a> SC

ft 4

3

Ia

—, .

#

•

• ^ - ^ .

o n

• o

0 o ° ° °

o o

Right Bank

(Ibi River)

J I I I I I L -2 0 2 4 6 8 Distance from the Mouth

Fig. 7 Run-up Height along the Ibi and Nagara Rivers

314
m 8.0

6.0

xiM

•H0

EC

? 4.0

c

3

03

Berm Width

A

X

5 m

2.0 -

_l_ 1.5 2.0 2.5 3.0 3.5 4.0

Wave Height

4.5 5.0m

Fig. 8 Effect of Berm Width on Run-up

V T.R +4.5m Berm

T.R + 3.8

1

T.R + 7.5 m

T.R +0m

Fig. 9 Cross Section of a Dike at the Mouth of the Ihi River

Embankment

Sumida River

3. )tZ

Land Fill

y///////////////////////^

Concrete Wall

Fig. 10 Cross Section of an Improved Dike at the Sumida River in Tokyo

315
NEW STRUCTURAL DESIGN ENFORCEMENT FOR BUILDING STRUCTURES

by

Kiyoshi Nakano

Makoto Watabe

Yuji Ishiyama

ABSTRACT

This paper summarizes the draft of a new aseismic design method for the Building

Standard Law Enforcement Order and related regulations which will be amended in the near

future. This aseismic design method aims at two targets: (i) the buildings shall withstand almost no damage subjected to the moderate earthquake motions which would occur

several times during the expected life of the buildings; (ii) the building shall not collapse nor harm human lifes from severe earthquake motions which would occur less than once

during the expected life of the building.

316
1 . GENERAL

1 . 1 Purpose

The purpose of this aseismic design method is to ensure that buildings shall withstand

moderate earthquake motions with almost no damage which might occur several times during

the expected life of the buildings, and shall not collapse nor harm human lives by severe

earthquake motions which might occur less than once during the expected life of the

buildings

.

1 . 2 Scope

This aseismic design method shall apply to buildings smaller than 60 meters.

2. DESIGN PROCEDURE

2. 1 Design for Moderate Earthquake Motions

(i) The stresses caused by lateral seismic shear prescribed in 3.1, 3.2 and 3.3 shall

not exceed the allowable stresses for temporary loads.

(ii) Each story drift of the building caused by the lateral seismic shear prescribed in

3.1 shall not exceed 1/200 of the story height. This value can be increased to 1/120 in

cases where non-structural members shall not experience severe damage under the increased

story drift.

2 . 2 Design for Severe Earthquake Motions

The ultimate lateral shear strength of each story shall not be less than the required

lateral shear, Qr, determined in accordance with the following formula.

QR

= DS• F ES • Q (1)

where, Dg = the structural coefficient which can be determined by the ductility and and

the damping, and shall be not less than 0.3. (See A-l)

Fgg = the shape factor which shall be the product of Fg and Fg. Where, Fg and Fg

shall be the factors which can be determined by the eccentricity of the

center of stiffness from the center of gravity and by the variation of the

lateral stiffness from the mean lateral stiffness, respectively, and both

shall be not less than 1.0. (See A-2)

Q = the lateral seismic shear for severe earthquake motions prescribed in 3.1.

317
2.3 Exceptions

(i) The design procedure of 2.1 (ii) and 2.2 do not apply to buildings listed in

Table 1.

(ii) The design procedure of 2.2 do not apply to buildings listed in Table 2.

3. LATERAL SEISMIC SHEAR

3. 1 Lateral Seismic Shear Above Ground Level

The lateral seismic shear, Q of each story above ground level shall be determined

using the following formula.

Q = Ci • W]; (2)

where, Cj = the lateral seismic coefficient of the I-th story as determined using

Formula (3).

Wj = the weight of the building above the I-th story.

The weight of the building shall be the sum of the dead load and the applicable portion

of the live load. In heavy snow districts, the effect of snow loads shall be considered.

The lateral seismic shear coefficient of the I-th story, Cx , shall be determined using

the following formula.

Ci = Z • RT • A i • Co (3)

where, Z = the seismic hazard zoning coefficient as shown in Figure 1.

Rj = the design spectral coefficient, which shall be determined by the type of

soil profile and the fundamental natural period of the buildings, as illustrated in Figure 2.

Ax = the lateral shear distribution factor, which shall be determined by the

fundamental natural period and the weight distribution of the buildings,

as shown in Figure 3.

Cq = the standard shear coefficient, which shall be not less than 0.2 for moderate

earthquake motions and 1.0 for severe earthquake motions.

The fundamental period of the building, T, to determine the design spectral coefficient

and the lateral shear distribution factor, shall be determined using one of the following

formulae.

313
T = 0.028 H for steel frame buildings. (4)

T = 0.020 H for other buildings. (5)

where, T = the fundamental natural period of the building in seconds.

H = the height of the building in meters.

3 . 2 Lateral Seismic Shear of Appendages

The lateral seismic shear of the penthouses, chimneys, towers, cisterns, parapets and

other appendages on buildings, q, shall be determined using the following formula.

q = k • w (6)

where, q = the lateral seismic shear of the appendage.

k = the seismic design coefficient of appendages and shall be 1.0, but the value

can be minimized to 0.5 when no harm will occur,

w = the weight of the appendage.

3.3 Lateral Seismic Shear of the Basement

The lateral seismic shear of basement, Qg, shall be determined using the following

formula.

Q B

= Qp + K • W B (7)

where, Qp = the portion of the lateral seismic shear of the first story that will act to

the basement.

K = the seismic design coefficient of the basement as determined using Formula

(8).

Wg = the weight of the basement.

The seismic design coefficient of a basement, K, shall be determined using the following

formula.

K = 0.5(1 - H_)Z«C (8)

40 °

where, H = the depth of the basement in meters and 20 meters in cases where the depth

exceeds 20 meters.

Z and C0 are the same as defined in 3.1.

319
ACKNOWLEDGMENT

The new aseismic design method proposed here depends mainly on "A Proposal for

Earthquake Resistant Method" which was carried out during 1972-1977 as the first national

project of the Ministry of Construction, Japanese Government. The authors wish to express

their deepest appreciation to all who helped with this project and to the members of the

aseismic design research committee in the Building Research Institute for their enthusiastic

discussion in drafting this design method.

It should be stated that a similar paper was submitted to 7th World Conference on

Earthquake Engineering, 1980, Turkey.

A-l THE STRUCTURAL COEFFICIENT, D S

1. The structural coefficient, Dg, can be determined using the following formula, in cases

where the damping ratio and the ductility factor is available from structural experiments

and analyses.

where, g =

/2u - 1

1.5

(A-l.l)

1 + lOh

u = the ductility factor

h = the damping ratio, and the standsrd value is 0.05 for reinforced concrete

buildings and 0.10 for steel encased reinforced and prestressed concrete

buildings.

2. In case the damping ratio and the ductility factor are not available, the structural

coefficient, Dg, can be determined based on standard values in the following table.

The value of Dg Types of Structures

0.3 - 0.4

0.4 - 0.5

0.5 - 0.75

0.75 - 1.0

Moment frame structures with excellent ductility

Structures with ductile shear walls or braces

Structures with shear walls or braces with poor ductility

Structures with very poor ductility

320
A-2 SHAPE FACTOR, FES

The shape factor, F E g

, of each story can be determined using the following formula.

FES = FE

FS (A-2.1)

where, FE can be given by the following table as a function of eccentricity of stiffness,

RE , given by the following formula.

r = e E —r

e

(A-2. 2)

where, e = the eccentricity of the center of the stiffness from the center of gravity

r e

= the elastic radius which can be defined as the square root of the torsional

stiffness divided by the lateral stiffness.

Table A-2.1 Shape Factor, FE , by Eccentricity of Stiffness, RE

REF

E

less than 0.15 1.0

0.15 < RE

< 0.3 linear interpolation

more than 0.3 1.5

And Fg can be given by the following table as a function of variation of lateral stiffness,

Rg, given by the following formula.

R

s = - b r (A-2. 3)

where, r = the lateral stiffness which shall be defined as the value of the story

height divided by the story drift caused by the lateral seismic shear for

moderate earthquake motions prescribed in 3.1.

r = the mean lateral stiffness which shall be defined as the arithmetic mean of

r's above ground level.

321
Table A-2.2 Shape Factor, Fg , by Variation of Lateral Stiffness, Rg

Reo F

S

more than 0.6 1.0

0.3 < Rs < 0.6 linear interpolation

less than 0.3 1.5

322
Table 1. Buildings Which Need Not Execute Design Procedure 2.1(ii) and 2.2

1 Buildings of conventional wooden construction, reinforced concrete block,

construction, or reinforced concrete panel construction that shall meet the

structural requirements stipulated to the relevant regulations.

2 Buildings not exceeding five stories of reinforced concrete wallconstruction, or reinforced concrete precast wall construction that shall

meet the structural requirements stipulated by the relevant regulations.

3

Buildings not exceeding 20 meters in height of reinforced concrete

construction, prestressed concrete construction, or steel encased reinforced concrete construction. Each story shall meet the following formula

25*^ + 7.AC

> Z«Wj (T-l.l)

where, ^ = the sum of the horizontal cross-sectional areas in square

centimeters of reinforced concrete shear walls in the

direction concerned.

Ac = the sum of horizontal cross-sectional areas in square

centimeters of columns.

Z = the seismic hazard zoning coefficient as shown in

Figure 1

.

W-j- = the weight in kilograms of the building above the story

concerned.

For steel encased reinforced concrete buildings, the coefficient 7, for Ac in the above formula should be increased up to 10.

323
Table 2. Buildings Which Need Not Execute Design Procedure 2.2

Buildings not exceeding 31 meters in height of reinforced concrete construction, steel

encased reinforced concrete construction, or steel construction that shall meet all of

the following requirements.

Each story above the ground level shall meet the following formula,

r > 0.6 (T-2.1)

where, r = the lateral stiffness which shall be defined as the value of the

story height divided by the story drift caused by lateral seismic

shear of moderate earthquake motions prescribed in 3.1.

r = the mean lateral stiffness which shall be defined as the

arithmetic mean of r's above ground level.

Eccentricity of the center of the stiffness from the center of gravity shall be

less than 0.15 of the elastic radius at each story. Where the elastic radius

shall be the square root of the torsional stiffness divided by the lateral

stiffness.

3a

Each story shall meet the following formula in cases of reinforced concrete

construction, or steel encased reinforced concrete construction.

25(AW + Ac

) > A: Z Wj (T-2.2)

= the sum of

of reinfor

Ac = the sum of

columns

.

Al = the latera

Z = the seismi

w

l = the weight

3b

Each story which has braces shall meet the following formulae.

o a

> (1 + 0.7 B)aB (T-2.3)

where, o a = the allowable stress for temporary loads.

3 = the ratio of the lateral shear of braces to the total lateral seismic

shear of the story.

Og = the stress of braces caused by the lateral seismic shear of moderate

earthquake motions prescribed in 3.1.

jPu > 1.2 MPy

where, J^u = the ultimate strength of the joint of the brace.

M^y = the yield strength of the brace.

(T-2.4)

324
Fig. 1 Seismic Hazard Zoning Coefficient, Z

325
Fig. 2 Design Spectral Coefficient, R

.326
327
IMPACT OF DRAFT HIGHWAY BRIDGE SEISMIC DESIGN GUIDELINES

by

James D. Cooper

Ronald L. Mayes

Roland L. Sharpe

ABSTRACT

A comprehensive set of draft seismic design guidelines for highway bridges has been

under development since 1977. The draft guidelines were completed in 1979 and have been

the center of a concerted evaluation effort by consulting engineering firms and State

departments of transportation. This paper presents the general philosophy behind the guidelines and describes the evaluation and redesign phase of the endeavor. Twenty-one bridges,

which were redesigned using the draft guidelines, are documented. Economic impact and

general viability of the draft guidelines along with recommendations for revisions to them

are presented.

328
INTRODUCTION

The 1971 San Fernando earthquake presented a major turning point in the development of

seismic design criteria for bridges in the United States. Prior to 1971 the American Association of State Highway Transportation Officials (AASHTO) specifications for the seismic

design of bridges were based in part on the lateral force requirements for buildings

developed by the Structural Engineers Association of California. [1] In 1973 the California

Department of Transportation [2] (CalTrans) introduced new seismic design criteria for

bridges that included the relationship of the site to active faults, the seismic response of

the soils at the site and the dynamic response characteristics of the bridge. In 1975

AASHTO adopted Interim Specifications which were a slightly modified version of the 1973

CalTrans provisions and made them applicable to all regions of the United States. In addition to these code changes the 1971 San Fernando earthquake also stimulated research activity on the seismic problems related to bridges. By 1977, significant earthquake engineering research studies relating to highway bridges had been completed and the Federal Highway

Administration (FHWA) funded a program with the Applied Technology Council (ATC) to develop

recommended new and improved bridge seismic design guidelines and then evaluate the impact

of these guidelines on the basis of design, construction, and cost implications. "Draft

Seismic Design Guidelines for Highway Bridges" [3] were completed in 1979 and based on the

evaluation phase of the study, final recommended guidelines will be prepared.

APPLICABILITY OF GUIDELINES TO THE UNITED STATES

A basic premise in developing the bridge seismic design guidelines was that they be

applicable to all parts of the United States. The seismic risk varies from very small to

rather high across the country. Therefore, for purposes of design, criteria were developed

for four seismic performance categories to which bridges would be assigned based on the

seismicity of the area in which the site is located and the bridge importance classification

(IC).

Bridges are classified according to their relative importance —either as an essential

bridge or not. An IC coefficient of II is assigned for essential bridges and I for all

others. Essential bridges are determined based on their social/survival and security/defense

classification. Essential bridges are those that must keep functioning during and after an

329
earthquake. Another aspect of classification covers the differing degrees of complexity and

sophistication of seismic analysis and design which are specified. This is called Seismic

Performance Category (SPC). SPC-D bridges include those designed for the highest level of

seismic performance with particular attention to methods of analysis, design and quality

assurance. SPC-C bridges include those where a slightly lower level of seismic performance

is required but the potential for damage is slightly greater than SPC-D. SPC-B bridges

include those where a lesser level of seismic performance is required and a minimum level of

analysis and specific attention to support design details are provided. SPC-A bridges

include those where no seismic analysis is required but attention to certain design details

for superstructure support is provided.

DESIGN PHILOSOPHY

The primary basis for development of the seismic design guidelines for bridges is to

minimize the hazard to life and provide the capability for bridges to survive during and

after an earthquake with essential bridges to remain functional. To meet this philosophy,

certain principles were followed:

Small-to-moderate earthquakes should be resisted within the elastic range of

the structural components without damage. Realistic seismic ground motion intensities should be used in the design procedure. Exposure to shaking from large

earthquakes should not cause collapse of all or part of the bridge. Where

possible, damage that does occur should be readily detectable and accessible for

evaluation and repair.

Conceptually there are two different approaches that are currently in use to satisfy

the above principles. These are mainly "force design" approaches and are the current New

Zealand [4] and CalTrans [5] approaches.

In assessing bridge failures of past earthquakes in Alaska, California, and Japan many

of the "loss of span" type failures are attributed in part to relative displacement effects.

Relative displacements arise from out of phase motion of different parts of a bridge, from

lateral displacement and/or rotation of the foundations and differential displacements of

abutments. Therefore in developing the draft guidelines the design displacements were considered to be just as important as design forces and for SPC-C and -D bridges, requirements

330
for ties between non-continuous segments of a bridge are specified in addition to minimum

bearing support lengths at abutments, columns and hinge seats.

The methodology used in the draft guidelines is in part a combination of the CalTrans

and New Zealand "force design" approaches but it also addresses the relative displacement

problem. The methodology varies in complexity as the SPC increases from A to D. Three

additional concepts are included in the draft guidelines that are not included in either

the New Zealand or CalTrans approach. First, minimum requirements are specified for bearing

support lengths of girders at abutments, columns, and hinge seats to account for some of

the important relative displacement effects that cannot be calculated by current state-ofthe-art methods. A somewhat similar requirement is included in the latest Japanese bridge

code. Second, member design forces are calculated to account for the directional uncertainty of earthquake motions and the simultaneous occurrence of earthquake forces in two

perpendicular horizontal directions. Third, design requirements and forces for foundations

are intended to minimize damage since most damage that might occur will not be readily

detectable.

For SPC-A bridges the only design requirement is one of providing minimal bearing

support lengths for girders at abutments, columns, and expansion joints. Even though the

level of seismic risk of these bridges is very small, prevention of superstructure collapse

was deemed necessary and hence the requirement. Design for the level of seismic force in

these regions was not considered necessary.

For SPC-B bridges the elastic member forces are determined by a single mode spectral

approach. Design forces for each component are obtained by dividing these elastic forces by

a reduction factor (R). For connections at abutments, columns and expansion joints, the

R-f actor is either 0.8 or 1.0 and they are therefore designed for the expected or greater

than the expected elastic forces. Foundations are also designed for the elastic forces.

For columns and piers the R-factor varies between 2 and 6 and they are therefore designed

for forces lower than expected from an elastic analysis and are expected to yield when

subjected to the forces of the design earthquake. Design requirements to ensure reasonable

ductility capacity of columns in SPC-B are not specified whereas they are for SPC-C and -D

bridges

.

For SPC-C and -D bridges the general approach is similar to SPC-B; however several

additional requirements are included. For columns, additional requirements are included to

331
ensure that they are capable of developing reasonable ductility capacity. For connections

and foundations alternate design forces to those determined by the procedures of SPC-B are

also permitted. These are based on the maximum shears and moments that can be developed by

column yielding when the bridge is subjected to the design earthquake forces. Horizontal

linkage and tie down requirements at connections are also provided. For SPC-D bridges,

settlement slabs are required to reduce the chance of abutment backfill settlement.

SEISMIC GROUND MOTION INTENSITIES

The selection of ground motion intensities to be used with the seismic design

provisions was carefully reviewed. Considerable study and effort had recently been made to

develop seismic risk maps and associated design spectra for the "Tentative Provisions for

the Development of Seismic Regulations for Buildings" (ATC-3-06) [6]. The ATC-3-06 maps

are based on (1) a realistic appraisal of expected ground motion intensities, (2) the probability that the design ground shaking will be exceeded is approximately the same in all

parts of the United States, and (3) frequency of occurrence of earthquakes in various

regions of the country. It is possible that the design earthquake ground shaking might be

exceeded, although the probability of this happening is quite small.

Ground motion is characterized by the use of two parameters - Effective Peak

Acceleration (EPA), Aa and Effective Peak Velocity-Related Acceleration (EPV), Av . Although

these parameters do not at present have precise definitions in physical terms, they should

be considered as normalizing factors for the construction of smoothed elastic response

spectra [7] for ground motions of normal duration. The EPA is proportional to spectral

ordinates for periods in the range of 0.1 to 0.5 seconds, while the EPV is proportional to

spectral ordinates at a period of about 1 second. [8] The constant of proportionality

(for a 5 percent damped spectrum) is set at a standard value of 2.5. Thus when the ordinates of a smoothed spectrum between the periods mentioned above are divided by 2.5, the

EPA and EPV are obtained. The EPA and EPV thus obtained are related to peak ground

acceleration and peak ground velocity but are not necessarily the same or even proportional

to peak acceleration and velocity.

3 32
METHODS OF ANALYSIS

The draft guidelines provide for 3 methods of analysis which vary according to the

refinement in the mathematical idealizat ion. They are the seismic coefficient method, the

single mode spectral method, and the multimode spectral method. All methods assume

simultaneous ground motion at all supports.

Simple Seismic Coefficient Approach - Method 1

This method does not require a calculation of the period of the bridge. In effect, the

method is designed to specify a lateral (i.e. , longitudinal or transverse) force equal to

some fraction of the bridge weight. The bridge, or bridge section, is designed to resist

the lateral seismic load applied along the center of mass of the superstructure.

Single Mode Spectral Approach - Method 2

This method requires a calculation of a period. The method is based on the premise

that the mode shape of the vibrating structure can be assumed and represented by a shape

function which can be expressed mathematically in terms of a single generalized coordinate

taken as the amplitude at the point of maximum displacement. It is designed to approximate

the dynamic character of the bridge. The method consists basically of the following steps:

(1) Determine the period for the assumed mode.

(2) Determine the corresponding seismic coefficient as a function of Av , structure

period, and soil type.

(3) Determine the maximum displacement due to the seismic loading.

(4) Determine the component forces corresponding to the maximum displacement.

Multimode Spectral Approach - Method 3

Bridges having complex geometry due to horizontal and vertical alignments and/or skew

support conditions have coupling in the three coordinate directions within each mode of

vibration. These coupling effects make it difficult to categorize the modes into simple

longitudinal or transverse modes of vibration. In addition, several modes of vibration

will in general contribute to the total response of the structure. These complex coupling

effects and multimodal contributions to the final response necessitate the use of computer

programs with space frame analysis capabilities for this method. Support motions applied in

33 3
any one of the two horizontal directions will generally yield forces about both principal

axes of the individual members due to these coupling effects.

BRIDGE REDESIGNS

Five State departments of transportation (California, Idaho, New York, Oklahoma, and

Washington) and four design consultants participated in the redesign and evaluation of the

"Draft Seismic Design Guidelines for Highway Bridges" [2] developed by the ATC Project

Engineering Panel. A total of 21 bridges were seismically redesigned using the draft guidelines. All bridge loadings remained the same for the redesigns except for the seismic loads.

The States redesigned their bridges based on the actual location of the structure. Thus the

SPC and analysis method was fixed by the guidelines. The consultants redesigned actual

bridges for all Seismic Performance Categories and related methods of analysis. A description of the bridges along with parameters used in the redesigns are presented in Table 1.

IMPACT OF DRAFT SEISMIC DESIGN GUIDELINES

Evaluation and impact of the draft seismic design guidelines is based on comments

received from the bridge designers who redesigned the 21 bridges listed in Table 1. It is

noted that the experience of the designers varied from those having little or no seismic

design time to those who routinely design bridges to resist earthquake imposed loading.

Some designers, particularly those who have little experience, tended to make the most

conservative interpretation of the guidelines, thus significantly increasing costs. The

first 5 bridges in Table 1 were redesigned by State bridge designers. The bridges were

originally designed using CalTrans criteria. The designers assumed that the value of Aa may

be as high as 0.7 in some parts of the State and thus used that value. In areas where Aa

exceeds 0.4, the draft guidelines recommend a maximum value of 0.4 be used in analyses.

The impact on total cost is seen to be minimal, varying from no change to an increase of

4.3 percent in structure cost. The designer for the fifth bridge assumed an A a of 0.4

which actually resulted in a 0.6 percent decrease in structural cost.

Bridges 6, 7, 8, 9, 14, and 15 were redesigned by several consulting engineering firms.

The respective bridges were redesigned as though they were sited in each of the SPC locations (A through D) and then those new details required by the guidelines were compared to

the actual design details. The fact that many of the bridges had no seismic design

334
provisions, coupled with conservative interpretations of the guidelines account for cost

increases of up to 55 percent. Typically, most of the increase was due to assuring that

foundation uplift did not occur. The cost increases would have been greatly reduced had

designers used ultimate capacities for piles and soils.

Bridges 7 and 8 probably give a more reasonable indication of the impact of imposing

more stringent design requirements in seismic areas of the country. Costs increased by 12

percent and 25 percent respectively when increasing design requirements for structures in

SPC-B to SPC-D. Increased foundation requirements account for most of the increased costs.

However, once again, utilizing ultimate capacities of piles and soil to resist earthquake

imposed loads would reduce these cost increases.

The remaining structures were redesigned by individual State departments of

transportation. Cost impact varies from no increase to 19 percent, again somewhat misleading

because of foundation overdesign.

REVISIONS TO DRAFT SEISMIC DESIGN GUIDELINES

The draft guidelines represent a radical departure from previous earthquake design

guidelines used in the United States. The redesign phase of the guideline development

effort has been beneficial in providing input which can be used as the basis for making

some apparent needed revisions to the guidelines. Most designers found the format of the

guidelines to be cumbersome. Consequently, it was recommended the format be changed.

Technical changes include the following:

(1) Add requirements for steel bridges. Specifically add a section to determine the

plastic forces for yielding of steel columns and add a section to assure that the columns

develop the required ductility capacity.

(2) Reduce the gap requirement (D*) at expansion joints. Redesigners indicated that

the gaps required by the guidelines were excessive. It seems reasonable to provide adequate

bearing support lengths to avoid loss of span and decrease the gap requirement and accept

some damage due to slamming.

(3) Recommend that bridges designed in SPC-B use Analysis Method 2 instead of 1.

Analysis Method 1 produces overly conservative design forces. The redesign results indicate

that seismic forces determined from Analysis Method 1 are greater than wind forces.

335
(4) Response Modification Factors have not been specified for pile bent requirements.

Specify factors for incorporation into the final guidelines.

(5) Specify to what force level column connections to bent caps should be designed.

(6) Clarify that the ultimate capacity of soils and piles should be used with the

foundation design forces. Give guidelines or references to determine the ultimate capacities. For SPC-B recommend a simplified method consistent with the design philosophy to

reduce the foundation design forces from the elastic forces.

(7) Consider if and when uplift of foundations can be permitted in the design of

foundations

.

In general, the guidelines were found to be realistic and workable. Most designers

stated that the guidelines represent a significant improvement over current criteria used in

the United States. Further, the guidelines are sufficiently general. Rail, mass transit,

as well as highway, organizations are reviewing the guidelines for possible use in the

earthquake design of their respective structures.

CONCLUSION

The draft seismic design guidelines developed by the Applied Technology Council for the

Federal Highway Administration represent a national earthquake design criteria for highway

bridges. Use of the guidelines, when modified to incorporate designer's recommendations

made during the bridge redesign phase of the study, will assure improved seismic resistance

for new bridges. Further, the increased seismic protection can be achieved for relatively

modest cost increase in parts of the United States where seismic design is not normally

considered.

ACKNOWLEDGMENTS

The guidelines presented in this paper are the result of a concerted effort by the

members of the ATC Project Engineering Panel: G. Fox, (Partner, Howard, Needles, Tammen and

Bergendorf), J. H. Gates (California Department of Transportation), V. M. Goins (Oklahoma

Department of Transportation), W. J. Hall (Professor, University of Illinois), E. V.

Hourigan (New York State Department of Transportation), R. Jarvis (Idaho Department of

Transportation), R. Kealey (Partner, Modjeski and Masters), J. R. Libby (Partner, James

R. Libby and Associates), G. R. Martin (Senior Engineer, Fugro, Inc.), J. P. Nicoletti

336
(ATC Board Representative and Senior Vice President, URS/Blume Engineers), J. Penzien

(Professor, University of California, Berkeley), W. Podolny (Federal Highway Administration),

R. H. Scanlan (Professor, Princeton University), J. D. Cooper (Project Manager, Federal

Highway Administration), R. L. Mayes (ATC Project Technical Director), and R. L. Sharpe

(ATC Project Director). Two project consultants have also made material contributions:

R. A. Imbsen (Engineering Computer Corporation) and D. Elms (Professor, University of

Canterbury, New Zealand).

The 21 bridge redesigns discussed in this paper were performed by the following

consulting engineering firms and State departments of transportation: Modjeski and Masters;

DeLeuw Cather; Howard, Needles, Tammen and Bergendorf; Moffat and Nichol, Idaho, Oklahoma,

New York, Washington, and California.

337
REFERENCES

Structural Engineers Association of California, "Recommended Lateral Force Requirements

and Commentary," Second Edition 1966.

Gates, J. H. , "California's Seismic Design Criteria for Bridges," Journal ASCE

Structural Division, December 1976.

Cooper, J. D. , Scheffey, C. F. , Sharpe, R. L. , and Mayes, R. L. , "Draft Seismic Design

Guidelines for Highway Bridges," Eleventh Joint Conference of the United States-Japan

Cooperative Program in Natural Resources, Panel on Wind and Seismic Effects, Tokyo,

Japan, September 1979.

Chapman, H. , "An Overview of the State of Practice in Earthquake Resistant Design of

Bridges in New Zealand," Proceedings of a Workshop on the Earthquake Resistance of

Highway Bridges, Applied Technology Council, Palo Alto, California, January 1979.

Gates, J. H. , "Factors Considered in the Development of the California Seismic Design

Criteria for Bridges," Proceedings of a Workshop of the Earthquake Resistance of Highway Bridges, Applied Technology Council, Palo Alto, California, January 1979.

Applied Technology Council, "Tentative Provisions for the Development of Seismic

Regulations for Buildings," prepared for the National Bureau of Standards/National

Science Foundation, ATC Report No. 3-06, Palo Alto, California, June 1978.

Newmark, N. M. , and Hall, W. J., "Seismic Design Criteria for Nuclear Reactor

Facilities," Proc. 4th WCEE, Santiago, Chile, 1969.

McGuire, R. K. , "Seismic Structural Response Risk Analysis, Incorporating Peak Response

Progressions on Earthquake Magnitude and Distance," Report R74-51, Dept. of Civil

Engineering, Mass. Institute of Technology, Cambridge, Mass., 1975.

338
Tab le 1

Summary of Bridge Redesigns Using Draft Seismic Design Guidelines for Highways Bridges

Bridge Name Length No. of \ * Analysis

f, Location Meters Spans SPC Molhod Superst ructure Substructure Foundation Abutments Comments

1. Kem River 123 3 0 .

7

c CalTrans Composite concrete Wall Piers Spread foun- Cantilever 3.31 increase in pier costs California c 1 slab on steel plate dation seat type 0.034 increase in founda- girders 19m width abutment tion costs

0.254 increase in super- structure/pier connection

Increases based on going

from CalTrans method to ATC-6 with Aa-0.7.

2. Route 80 on 212 7 0.7 c 2 2Frames-CIP/PS Box Single cantilever Concrete cast Expansion typeil.S% increase in column

Ramp 0.2 c Girder and RC Box columns in drilled with end dia- costs California Girder 12m wide holes phragm an pile: 1.0% increase in foundation

0.31 increase in restrainers

1.5% increase in hinges

Increases based on going from CalTrans method to ATC-6 with Aa

»0. 7.

• 3-Milliken Ave. 139 7 1 0.4 c Reinforced concrete 2 column bents Concrete foot- Cantilever Redesign not completed. California 1 0.2 c 2 »x girders 13m wide ings on CIDH type with back-Designer states that results piles wall from method 2 of analysis

- may be on conservative side.

4. Verdugo Blvd. 160 5 0.7 c 3 UP reinforced box Multi-column Spread footing sDiaphragm type! No changes required compared

Cali fomia Av=0 .

4

c 3 and short stee . on spread to original design. H piles footings

S.Camden Ave. 69 2 0.4 c 1 Composite concrete 2 column bent Spread footing ; Monolithic 0.3% increase in column

California 0.3 c 1 slab on steel box abutment on costs girders 21m wide spread footing 0.5% increase in footing

costs

1.4% decrease in abutment

costs Decrease of 0.6% based on going from CalTrans method

to ATC-6 with Aa=0.4.

6. Hudson River 109 5 0.05 Continuous rolled 3 Column Re inf. Pile Support- Stub Abutment

f

Cost increases were incurred

Valley 0.1 B 1 beam - concrete Cone. Bents Squ. ed Concrete w/wind walls in columns, footings, and New York 072" c 1 structure 11m wide Columns w/pier Slabs supported by abutments as the force level 0.3 c 1 caps piles increased.

0.4 D 2 Designed by 1957 AASHTO.

7-Humbolt River 88 3 0.05 A Cast in place cone. Four Column Pile footing Open end seat Costs varied by 12% between

Nevada o.i B 1 box girder 17m wide Bents. Skew varie with spread SPC-B and SPC-D designs with

0.2 C 2 w/396m radius of between 30 to 40 footings 70% in the abutment.

0.3 C 2 curvature degrees Designed by 1977 AASIffO. 0.4 D 2

8. Durham Road 62 2 0.05 A Cast in place cone. Four Column Pile footing Monolithic w/ Costs varied by 25% between

California 0.1 B 1 box girder 26m wide bent deck on pile j SPC-B and SPC-D designs with

0.2 C 2 footings , 70% in the foundation. 0.3 C 2

i Designed by 1977 AASIITO & 0.4 D 2

\ i CalTrans criteria.

9. Menomonee Riv 73 2 0.1 B 1 Continuous welded | L^2m wide x 6m I Pile footing j 1-Free Stand- Significant cost increases Wisconsin i 0.2 C 2 steel composite , long pier with 1

! ing full in foundations as force 0.3 C 1 plate girder j 11 degree skew 1

; retaining, level increases. Total cost 0.4 D I j 9m wide 1 1-Free Stand- increased by 9-43%.

ing cellular. Designed by 1976 AASirTO. Both pile L—~

1 supported

10-Lewiston 77 3 0.05 A N/A Cast in place cone. 1 3 columns per Spread | Concrete cap No change in the design or

Idaho box girder 24m wide bent footings supported on 1 related costs. No seismic

j —1.

| piles design.

Underlined values indicate approximate initial design values used.

339
Table 1 (Continued)

Rntlge Name

r

i location

Length MetersNo. of Spans A

* n sre

Analysi s Method Supcrst ructurc Substructure 1 oundation Abutments Comments

11. Heybum

Idaho

400 32 0.10 V0.15

c 1 ^estressed T-beams

w/concrete deck

L3m wide

Steel shell pile

bents filled with concrete with

35° skew

Piles and

columns are continuous

Beam type supported by piles

A 194 increase in total cost

due to additional pile

requirements. Designed by 1975 AASHTO.

12. Verdigris Riv. Oklahoma

137 5 0.10 B

1 1

1 'restressed cone.

I beam 11m wide

2 columns w/pier

cap and web wall Spread footing Cantilever typi on piles with

wing walls

Slight increase (=24 of

total) in cost on one pier. Designed by 1976 AASHTO.

13. Poteau River Oklahoma

174 5 0.1 B 1

j

1

Reinforced concrete

girders on approach

spans . Haunched

steel plate girder

center 3 spans. 11m wide.

2 Columns per ben w/a web wall and

tie beam between

columns

.

t Two footing conditions were included. A an lied shaft and a spread footing

Cellular type

abutments. The approach spans form the top

part of the abutment

.

^ 3.64 (of total cost) increase in cost of each foundation. Designed by 1976 AASHTO.

H.Michelson Dr. California

107 4 0.05 0.10

K1—

0.2 0.3

0.4

A

Bc

c

D

|Precast-pres tressed

1 [concrete girders

2 modified L/reinforced concret

2 modified Ideck. 13m wide

2 modified

j

Solid reinforced

concrete walls

? 1 3m long 51cm thick

7m high

45 ton pre- stressed cone. Open end seat Additional construction

type, free 'cost increased by 24 for standing with \ SPC-A to 474 for SPC-D with

back wall ' 2/3 of the cost increase in the piers. Designed by 1973 ,AASrrm and Cal Trans.

i bourn t. srree

Cal ifornia

195 g 0.05

0.10

0.20

0"7T~ 0.4

A

Bc

c

n

Cast in place R.C.

3 T beams. 2 inter3 mediate hinges and

3 [expansion joints at

3 ;ea. abutment. 12m wid

1

Sol id reinforced

concrete piers

13m long 11m high

51cm inches thick

20° skew

50 ton driven

steel H piles Open end seat Construction cost increased

type free ! by It for SPC-A to 55% for standing with i SPC-D with 804 of the cost back wall i occuring in the pier foun- dation. Designed by 1973

: AASHTO and CalTrans.

16.Battenkill

New York

55 1 0.1 „ Anon anA can* An inrrOQCO 1 n trie met n f~ D 1 IT 1 C> L I [AJi L Pensioned concrete

beams with mem.

(concrete slab

14 m wide

type on spread

foundations

.

the abutment bearings. In the calcs a reduction of

2.5 was not used but is nermitted.npiionpri hv MYnrrr <;npr.

17. Rt. 225 New York

71 2 0.1 B 1 and 2 Steel plate girders w/concrete slab

19m wide

3 column bent Spread foot in ; Open end seat type on spreac foundat ions

Major redesign of bearings. Foundation size increased. Designed by NYDOT spec.

18. Ihilett Street

New York

192 3 0.1 B 1 and 2 Steel plate girders w/concrete slab

13m wide

Two column bent 35 ton cast

in place con- crete piles

Open seat abutments supported on piles

Large expense in increasing

the gap separation. Bearings

require redesigning. No change in column design.

Dps i pied bv NYDOT spec.

19. Rt. 28A New York 145 2 0.1 B 1 and 2 Steel plate girders w/composite slab

11m wide

Two column bent Spread foot- ing Open seat abutment on

spread foun- dat ions

No change in footing and

column design. Large gap separation at joint. Designed by NYDOT spec.

20. Midway Avenue Washington

73 2

1

j 0.2

1

C 1 Precast girders, continuous for live

load. 13m wide

Single column

7m high

! Spread foot-

| ings Cantilever

bearing wall

w/ spread

footings

Small cost increase (14) in footings and increasing gap

width. Designed by 1977 W DOT spec, for E.Q.

21. Yakima River Washington

174

1

6 0.10 IV, 0.15

2 Prestressed girder

1 16m wide. Two column bents ! Spread foot-

: uig- Stub wall An 84 increase in total cost compared to original structure. Designed by 1977 W -TOT

* Underlined values indicate approximate initial design values used.

340
NEW SPECIFICATIONS FOR EARTHQUAKE-RESISTANT

DESIGN OF JAPANESE HIGHWAY BRIDGES

by

Eiichi Kuribayashi

Toshio Iwasaki

Osamu Ueda

ABSTRACT

This paper briefly discusses the history of seismic design provisions for highway

bridges in Japan, and introduces the new specifications (Japan Road Association, 1980) for

earthquake-resistant highway bridge design.

341
INTRODUCTION

Since the Kanto Earthquake of 1923 Japan has experienced a number of severe

earthquakes, and the incidence of damage to highway bridges is considerable. Stemming from

the damage caused by the Kanto Earthquake, seismic forces were first taken into account in

highway bridges design in 1926. The seismic coefficient method in the practical design of

structures was developed and introduced at that time. After experiencing severe damage

during consecutive strong earthquakes, seismic regulations were reviewed and amended several

times. In view of the damage caused by the Niigata Earthquake of 1964 the amended specifications for earthquake-resistant design of highway bridges were issued in 1971 by the Japan

Road Association. Much work has been done to establish more rational seismic criteria for

highway bridges. This includes recent advancements in earthquake engineering associated

with bridges, the damage experience due to the Miyagi-ken-oki Earthquake of 1978, and the

new specifications which were completed in March, 1980. This paper briefly describes the

history of highway bridge seismic design codes in Japan, and introduces the new specifications (JRA, 1980) for highway bridge seismic design.

HISTORY OF EARTHQUAKE RESISTANT DESIGN PROVISIONS FOR HIGHWAY BRIDGES IN JAPAN

The Ministry of Home Affairs stipulated, in 1926, the "Specifications for Design of

Road," which are parts of the "Road Laws." In the specifications, seismic forces were

first taken into account in the design of highway bridges, since several highway bridges

sustained substantial damage during the 1923 Kanto Earthquake. The specifications provided

that highway bridges be designed using the seismic coefficient method, in which horizontal

seismic coefficients were taken from 0.1 to 0.4. The values of the coefficients were dependent on areas and ground conditions. For bridges to be constructed in Tokyo and Yokohama,

seismic coefficients of 0.3 or more were recommended. This seems due to the substantial

damage to bridge structures in the areas during the Kanto Earthquake.

The Ministry of Home Affairs issued, in 1939, "Specifications for Design of Steel

Highways," which took place of the earlier specifications. The new ones stipulated both a

horizontal coefficient of 0.2 and a vertical coefficient of 0.1.

The specifications were revised again in 1964 by the Japan Road Association, with a

commission from the Ministry of Construction. The revised specifications specified that

342
both a horizontal coefficient of 0.1 to 0.35 depending on areas and ground conditions, and

a vertical coefficient of 0.1 shall be considered in the aseismic design.

In view of extensive damage to bridge structures from the 1964 Niigata Earthquake, the

Japan Road Association, also with a commission from the Ministry of Construction, drew up

in January 1971, comprehensive specifications [1,2] exclusively for earthquake-resistant

design of highway bridges. In the 1971 specifications, two methods are provided for a seismic design. One is the conventional seismic coefficient method for rigid structures, where

the horizontal coefficient ranges between 0.1 and 0.24 depending on areas, ground conditions,

and importance. The other is the modified seismic coefficient which considers structural

responses for comparatively flexible structures, where horizontal seismic coefficients vary

from 0.05 to 0.3 depending on fundamental natural periods in addition to the three factors

above.

During these years after the 1971 specifications were issued, technological

advancements in bridge engineering and earthquake engineering have been remarkable. Especially comprehensive research works were executed in the New Aseismic Technology Development Project of the Ministry of Construction from 1972 through 1976. The results of the

investigations achieved in this project were put into a unified form of provisions on earthquake resistant design for civil engineering structures and building structures, and "A

Proposal for Earthquake Resistant Design Methods," was issued by the Ministry of Construction in March, 1977 [3,4], Also, the Miyagi-ken-oki Earthquake of June 12, 1978 caused

extensive damage to numerous bridges [5]. In view of the results of the above investigations and the damage features, the Japan Road Association amended the 1971 specifications

and stipulated, in March 1980, new specifications for earthquake resistant design of highway

bridges [6]. The new specifications became a part of "Specifications of Highway Bridges,"

which consist of five parts, Part I General Specifications (1972), Part II Steel Bridges

(1972), Part III Concrete Bridges (1977), Part IV Substructures (1980), and Part V Earthquake Resistant Design. The new specifications (Part V Earthquake Resistant Design) will

be outlined below. [6,7,8]

Further, characteristic criteria was proposed, between 1966 and 1968, tentatively for

the aseismic design of highway bridges relating to specific projects administrated by the

Japan Highway Public Corporation (JHPC), the Metropolitan Expressway Public Corporation

(MEPC) , the Hanshin Expressway Public Corporation (HEPC) and the Honsyu Shikoku Bridge

343
Authority (HSBA). The Japanese National Railways (JNR) stipulated in 1968, and revised in

1979, its own criteria for aseismic design of railway bridges.

Table 1 lists briefly this history of design loads (primarily seismic loads) for

highway bridges in Japan. [9]

NEW SPECIFICATIONS (JRA-1980)

Outline

Efforts to revise the 1971 specifications into the form of Part V. of the

Specifications for Highway Bridges have been underway by the Earthquake Resistant Design

Subcommittee from 1977 to 1980, and the new specifications were issued in April, 1980. The

contents of the new specifications are presented in Table 2. They apply to the design of

highway bridges with span lengths not longer than 200 meters.

The specifications basically stipulate using seismic coefficient methods and provide

two methods in determining design seismic coefficients. One is the conventional seismic

coefficient method that applies to the design of relatively rigid structures. The other is

the modified seismic coefficient method which considers structural responses that apply to

the design of relatively flexible structures. Provisions on seismic motions in dynamic

analysis, and seismic coefficient in ductility analysis, were newly introduced.

The principal features and improvements of the design methodology in the new

specifications are described as follows:

Seismic Coefficient Method

(1) In the seismic coefficient method for relatively rigid structures, the horizontal

design seismic coefficient (kh) shall be determined by

kh

= vi*V2*V3«ko (1)

where kh = horizontal design seismic coefficient,

kg = standard horizontal design seismic coefficient (= 0.2),

\>l = seismic zone factor,

\>2 - ground condition factor,

U3 = importance factor.

344
The values of vj , V2 , and V3 are shown in Tables 3, 4 and 5, respectively. The minimum

values of shall be taken as 0.1.

(2) The vertical design seismic coefficient may generally be considered as zero,

except for special portions, such as bearing supports.

(3) The horizontal design seismic coefficient for structural parts, soils, and water

below the ground surface may be considered as zero.

(4) Hydrodynamic pressures and earth pressures during earthquakes are specified in the

specifications.

(5) Special attention is paid to very soft soil layers and soil layers vulnerable to

liquefaction during earthquakes. The bearing capacities of these layers were either

decreased or neglected in the design, in order to assure high earthquake-resistance for

structures that are built in these layers.

(6) Special attention is also paid to the design of structural details, because of

damage previously experienced on bridge structures. Provisions are specified for bearing

supports and devices to prevent bridge girders from falling.

(7) Increases in allowable stresses of materials may be considered in the

earthquake-resistant design, magnitudes of increases for various materials are specified

in several related specifications.

The increasing rates are as follows:

concrete in reinforced concrete structures: 50%

reinforcements in reinforced concrete structures: 50%

structural steel for superstructures: 70%

structural steel for substructures: 50%

concrete in prestressed concrete structures subjected

to compressive forces: 65%

foundation soils: 50%

Seismic Zoning Map

The newly developed seismic zoning map illustrated in Figure 1 was adopted. The map is

based on the Proposal for Earthquake Resistant Design Methods [3,4] is intended to unify

seismic zoning maps currently applied to civil engineering structures and buildings. Slight

modifications were introduced to the proposed original map out of administrative considera345
tions. The values of , are 1.0, 0.85 and 0.7 for A, B, and C zones, respectively, as

shown in Table 3.

Classification of Ground Conditions

In the previous specifications, the classification of ground conditions was determined

in accordance with geological conditions. However, since subsurface ground responses during

earthquakes would generally be more largely affected by the predominant period of the ground,

it is considered more reasonable to classify grounds into groups in terms of the period of

the ground. Consequently in the new specifications, the ground conditions are classified

into four groups according to Table 4, in which the characteristic value of ground, Tg, is

stipulated to be principally calculated by the following equation:

T = £4Hi (2) S

i Vsi

where Tg = Characteristic value of ground (second)

Hi = Thickness of i~th subsoil layer (m)

= Shear wave velocity of i-th subsoil layer at low strain (around 10

-

^ percent)

As for shear wave velocities, it is recommended that it be directly measured through

site investigation. Shear wave velocities may be assumed from N-values of standard penetration tests by

1/3

v = r 100 N (1 < N < 25) for cohesive soils

( 3 ) s 1 80 N

1/3 (1 < N < 50) for sandy soils

The baserock for calculation of Eq. (2) is stipulated to take on the soil layer that

has a shear wave velocity at low strain equal to 280 m/sec or higher and is not underlaid

by materials having significantly lower shear wave velocities.

The characteristic value Tg implies a natural period of subsurface ground at low strain

levels. The classification of Tg shown in Table 3 was proposed from numerical seismic analyses of various types of subsurface grounds. Such analyses revealed that the natural period

T s of subsurface ground at high strain levels which would be expected to occur during strong

earthquakes can be approximately obtained by the following equation.

T s = 1.25 T

g (4)

346
It was also found that ground conditions could be adequately classified into four

groups by taking T s as T s < 0.25 seconds, 0.25 < T s < 0.5 seconds, 0.5 < T s

< 0.75 seconds

and T s

> 0.75 seconds. The characteristic values Tg presented in Table 4 were thus obtained

by substituting the above mentioned T s into Eq. (4).

Figure 2 is one of representative results of analyses showing a relationship between the

characteristic values Tg and the thickness of soil deposits. It is apparent from the result

that the classification of ground conditions determined by the characteristic value Tg as

shown in Table 4 can also be approximately estimated by the thicknesses of alluvial and

diluvial layers. It is therefore recommended to use this relation to classify the ground

condition when Tg cannot be obtained.

Figure 3 shows a comparison of the ground classifications which were provided in the

1971 specifications and in the new specifications. It is understood from the results that

some ground areas which are evaluated as Groups 1 and 2 in the current specifications turn

into Groups 2 and 3, respectively, in the new specifications.

Liquefaction of Sandy Soil Layers

In the 1971 specifications, it was stipulated that saturated sandy soil layers that are

within 10 meters of the actual ground surface, and that have a standard penetration test

N-value less than 10, a coefficient of uniformity less than 6, and also a D20 -value on

the grain size accumulation curve between 0.04 mm and 0.5 mm, shall have a high potential

for liquefaction during earthquakes, and that bearing capacities of these layers shall be

neglected in design.

After the Niigata Earthquake, comprehensive studies were conducted to assess the

vulnerability of saturated sandy soils. Based on these studies, the provisions for liquefaction are improved in the new specifications as follows:

(1) Sandy Soil Layers Needed to be Checked for Liquefaction — Saturated alluvial

sandy layers which have a water table within 10 meters from the ground surface, and have

D5Q-values on the grain size accumulation curve between 0.02 and 2.0 mm, are vulnerable to

liquefaction for the depth between 0 and 20 m, and liquefaction potential of these layers

shall be estimated according to item (2).

347
(2) Estimation of Liquefaction — For those soil layers which are judged to be

vulnerable for liquefaction, liquefaction potential shall be checked based on liquefaction

resistance factor, F^, defined by the following equation.

F

T = » (5) L L

where F^ = liquefaction resistance factor

R = resistance of soil elements to dynamic loads, and

R = Rj + R 2 (6)

R^ and R2 shall be determined in accordance with Figures 6 and 7, respectively.

L = dynamic loads to soil elements induced by earthquake motion

L = rd ks i (7)

°v

r d

= 1.0 - 0.015z (8)

z = depth from the actual ground surface (m)

k s = seismic coefficient for evaluation of liquefaction, and shall be determined by

the following equation:

k s = Vi.v2-v3.kgo (9)

vl,v2,v3 = seismic zone factor, ground condition factor, and importance factor, provided

in Tables 3, 4 and 5, respectively.

k so

= 0.15

o ov

= total overburden pressure (kg/cm )

1 2 ov

= effective overburden pressure at the static condition (kg/cm )

Soil layers having a liquefaction resistance factor F^ smaller than 1.0 shall be judged

to liquefy during earthquakes. Figures 4 and 5 are graphic illustrations of the first term

R\ and the second term R2 represented in the following equations, which were proposed based

upon the results of laboratory dynamic triaxial tests on soil specimens taken from several

sites in Japan [10,11].

348
(0.02 mm < D

5Q

< 0.05 mm)

(0.05 mm < D

5Q

< 0.6 mm)

(0.6 mm < D

5Q

< 2.0 mm)

(3) Treatment of Soil Layers which were Judged to Liquefy — For those soil layers

which are judged to liquefy by the above estimation and are within 20 meters of the actual

ground surface, bearing capacities and other soil constants shall be either neglected or

reduced in the seismic design by multiplying the original bearing capacities by reduction

factors Dg which are determined in accordance with F^-values and tabulated in Table 6.

Modified Seismic Coefficient Method

In the 1971 specifications, the modified seismic coefficient method was provided to

apply to bridges which have flexible piers and fundamental periods longer than 0.5 seconds,

such as those with piers taller than 25 meters above the ground surface. Accounting for

seismic responses, magnification factors (3) for the modified seismic coefficient method

were stipulated. However, it has been pointed out that fundamental natural periods sometimes exceed 0.5 seconds even for those bridges with piers lower than 25 meters above the

ground surface. Using experimental data on the relationship between fundamental natural

periods and pier heights, it is modified in the new specifications so that the modified

seismic coefficient method shall apply to bridges which have flexible piers and long fundamental periods, such as those with piers higher than 15 meters above the ground surface.

In addition to the above change, the following two modifications were also introduced:

(1) The magnification factors (3) are modified as shown in Table 7 and Figure 6 so as

to avoid a sudden change of 3~value at a period of 0.5 seconds.

(2) In the 1971 specifications, the effects of subsoil conditions were not considered

in estimating fundamental natural periods. Since the effects of subsoils would be predominant in calculating fundamental natural periods, especially for bridges with short piers,

it is stipulated in the new specifications that the effects of subsoils shall be taken into

account for those bridges which are constructed in the soft ground. It is recommended that

the fundamental natural period for the individual system consisting of each substructure

349

0.0882 « / N + o.l9

V

R = ( 0.0882 L_ N + 0.225 lng 10 ( °- 35 ')

}l o'v + 0.7 D

50

0.0882 _ / N - 0.05 —1 Vo; + 0 .7
and the part of superstructures supported by it be estimated using the following

equation.

T = 2.01 /& (11)

where T = Fundamental natural period in seconds of the system consisting of a substructure

and the section of the superstructures supported by it.

6 = Maximum horizontal displacement (in meters) of the pier when subjected to the dead

weight of the section of superstructure supported by the substructure and also to

80 percent of the dead weight of the substructure above the ground surface assumed

in earthquake resistant design.

Seismic Motions in Dynamic Analysis

In the 1971 specifications, it is stipulated that dynamic earthquake response analyses

shall be adopted for those bridges for which detailed investigations are required. In the

new specifications, an article is introduced concerning the seismic motions to be utilized

in dynamic response analyses. The principal provisions are as follows:

(1) Dynamic response analyses may apply to those bridges which are designed either by

the seismic coefficient method or the modified seismic coefficient method, in order to

investigate precisely the earthquake resistivity of bridges in terms of ductilities and

maximum bearing capabilities. Dynamic analyses are needed for those bridges having structural systems which are significantly different from those assumed in the seismic coefficient method or the modified seismic coefficient method, those bridges having new structural

types for which the experience of damage accumulated from past earthquakes cannot be adequately extended, those bridges which are constructed on extremely soft soil deposits and

are expected to deform considerably during earthquakes, and those bridges for which detailed

investigations on requirements of ductility of structures are needed.

(2) Two types of dynamic earthquake response analyses, i.e., response spectrum

analyses and time history analyses can be used.

(3) Input motions used for the time history analyses shall be selected from

strong-motion acceleration records which consider the dynamic characteristics of bridges

and the characteristics of the records.

350
In determining input seismic motions, two procedures are proposed. One is to estimate

expected intensities at the side based on the life-time of the bridge and the recurrence

period of earthquake events. Another procedure is to estimate the expected ground motions

by assuming the locations and the magnitudes of specific earthquakes around the site. In

the second, ground motions can be evaluated either by the theory of seismic gaps or the

statistics of the past historical earthquakes. It is also recommended that the input seismic motions be selected according to the objectives of the earthquake response analyses.

It states that bridges shall maintain their functions for those motions which are expected

to occur two or three times during the bridge lifetime, and they shall survive those motions

which are expected to occur once, or rarely, at the site.

(4) In utilizing seismic ground motions recorded on soft soil deposits which have

appreciably different ground conditions compared to those at a planned construction site,

it is recommended such effects be taken into account in the analyses. For such purposes,

earthquake response analyses based on the baserock motions are recommended.

(5) Input earthquake response spectra used for the response spectrum analyses shall be

determined using the response spectra calculated from strong-motion accelerations. In an

appendix to the new specifications, the results of statistical analyses of strong-motion

acceleration records are presented. Some of the representative earthquake response spectra

and relations between maximum horizontal accelerations and epicentral distances are

presented in Figures 7 and 8, respectively [12].

Seismic Coefficient in Ductility Analysis

In order to avoid brittle failure during earthquakes, it is extremely important that

reinforced concrete structures have adequate ductility. A provision that stipulates the

seismic coefficient used for the design of reinforced concrete piers with ductility is

introduced. It stipulates that the seismic coefficient in ductility analysis shall be

determined using the following equation.

khd = V4 • kh

(12)

where k^d = seismic coefficient in ductility analysis

V4 = structural characteristics factor (greater than 1.3)

k

h

= horizontal design seismic coefficient provided in Eq. (1).

351
Table 8 shows maximum ductilities of ordinary RC bridge piers which were analytically

determined to account for deformation due to bending of piers and deformation of reinforcement pulled out from footings, in which the critical strains of concrete were assumed as

0.35 percent. It can be recognized from these results that the maximum ductilities of

bridge piers normally designed by the seismic coefficient method can be taken as approximately 6. However, since values for maximum ductilities are derived from analytical calculations for a half cycle loading, it is considered desirable to take maximum design ductilities to be smaller than 6. Considering the fact that concrete pier ductility decreases

significantly under alternatingly repeated loading conditions [13], one third of the values

tabulated in Table 8, which lead to about 2, is recommended as the ductility factor for

design purposes.

CONCLUSIONS

Earthquake resistant highway bridge design criteria in Japan is briefly described with

emphasis given to improvements and modifications in the new specifications (JRA-1980). In

view of the history of the earthquake resistant design of highway bridges, it is considered

necessary to concentrate a comprehensive investigation on the following subjects in the

future.

( 1 ) Analysis of the Effects of Soil-Structure Interactions

Bridge construction on deep, soft soil deposits has increased recently. From the

evidence of past, extensive earthquake damage, it is well recognized that the influence of

surrounding subsurface soils are very important for the seismic responses of substructures,

especially for substructures which are embedded in deep soft ground. Consequently, considerable interest has been focused on the soil-structure interaction effects on such structures by model experiments and theoretical analyses. However, very limited research has

been undertaken to investigate the seismic response of actual substructures during strong

seismic excitations. Investigations of the effects of soil-structure interaction which

utilize strong-motion records obtained at actual bridges are encouraged. For this purpose,

it is recommended that strong motion observations, especially simultaneous observations of

both the bridges and the surrounding subsurface grounds, be extended.

352
(2) Analysis of Seismic Behavior of Substructures in Liquefied Soil Layers

Due to the comprehensive research conducted after the 1964 Niigata Earthquake,

determining the vulnerability of saturated sandy deposits and judging in situ liquefaction

potential became both possible and practical. However, further investigation of the seismic

behavior of substructures in liquefied layers is needed, as is the development of suitable

earthquake resistant design procedures for bridges under such conditions.

(3) Experiments on Ductilities of Bridge Piers

Seismic damage to bridges were most commonly caused by pier and foundation failures.

It is, therefore, extremely important to prevent brittle failure to substructures. Up to

the present, only very limited experimental studies have been conducted on the hysteretic

behavior of bridge piers under cyclic loading. Such a lack of data on the hysteretic

response of piers is one of the major obstacles to introducing limit design to bridges to

account for the ductility of members. It is recommended that extensive efforts be devoted

to accumulating such experimental data.

(4) Design Details of Bearing Supports and Connections

For providing highway bridges with adequate resistance to seismic disturbances, it is

very important to give special attention to the design details of connections between superstructures and substructures, to preventing girder fall, and to avoid severe damage which

is caused by bearing support failures.

353
REFERENCES

[1] Japan Road Association, "Specifications for Earthquake Resistant Design of Highway

Bridges," 1971.

[2] Kawakami, K. , Kuribayashi, E., Iwasaki, T. and Iida, Y. , "On Specifications for

Earthquake Resistant Design of Highway Bridges," (January, 1971), 7th Joint Meeting,

U.S. -Japan Panel of Wind and Seismic Effects, UJNR, Tokyo, 1975.

[3] Ministry of Construction, "A Proposal for Earthquake Resistant Design Methods," 1977

(In Japanese).

[4] Nakano, K. and Ohashi , M. , "A Proposal for Earthquake Resistant Design Methods," 9th

Joint Meeting, U.S. -Japan Panel on Wind and Seismic Effects, UJNR, Tokyo, 1977.

[5] Okubo, T. and Iwasaki, T. , "Summary of Experimental and Analytical Seismic Research

Recently Performed on Highway Bridges," Workshop on Research Needs of Seismic Problems

Related to Bridges, Palo Alto, California, USA, 1979.

[6] Japan Road Association, "Specifications for Highway Bridges," Part V, Earthquake

Resistant Design, 1980 (In Japanese).

[7] Ohashi, M. , Kuribayashi, E., Iwasaki, T. , and Kawashima, K. , "An Overview of the State

of the Practice in Earthquake Resistant Design of Highway Bridges in Japan," Workshop

on Research Needs on Seismic Problems Related to Bridges, Palo Alto, California, USA,

1979.

[8] Kuribayashi, E., Iwasaki, T. , "Progress Report on Research Work in Bridge Earthquake

Engineering at the Public Works Research Institute, Japan," Seventh World Conference

on Earthquake Conference, Istanbul, Turkey, 1980 (to be presented).

[9] Iwasaki, T. , Penzien, J., and Clough, R. W. , "Literatury Survey - Seismic Effects on Highway Bridges," EERC Report No. 72-11, University of California, Berkeley, 1972.

[10] Ohashi, M. , Iwasaki, T. , Tatsuoka, F. and Tokida, K. , "A Practical Procedure for

Assessing Earthquake-Induced Liquefaction of Sandy Deposits," 10th Joint Meeting,

U.S. -Japan Panel on Wind and Seismic Effects, UJNR, Washington, D.C., 1978.

[11] Iwasaki, T. , Tatsuoka, F. , Tokida, K. and Yasuda, S., "A Practical Method for

Assessing Soil Liquefaction Potential Based on Case Studies at Various Sites in

Japan," Second International Conference on Microzonation for Safer Construction

Research and Application, San Francisco, 1978.

[12] Iwasaki, T., Katayama, T. , Kawashima, K. and Saeki, M. , "Statistical Analysis of

Strong-Motion Acceleration Records Obtained in Japan," 2nd International Conference on Microzonation for Safer Construction Research and Application, San Francisco, 1978.

[13] Osaka, Y. and Ohta, M. , "Experimental Study on the Behavior of Reinforced Concrete

Bridge Piers under Seismic Alternating Load," A. I. C.A.P.-C.E.P. Symposium on Structural Concrete under Seismic Actions, Rome, 1979.

354
Table 1 History of Design Loads for Highway Bridges in Japan

(Primarily Seismic Loads)

Design Live Loads

Seismic Loads

k: Horizontal Selaalc Other Seismic

Regulat lona

(k: aelamlc coef.) Year Kaae of Regulations Class

Truck

Roller

Major Earthquake

1) 180b Order No. 13 Ministry of Howe Affairs

(MHA)

U-454 kg/a not considered (M>7) after Rika

2) 1906 Order

MHA

R-13.6 t

S-12.

7

U-400 600 kg/m

(carriage way) U-270 400 kg/m

(footway)

not considered

(1971)

3) 1919 Road Laws

MHA- R-13.6 t T-11.2 l S-30 t

Sane as 2)

dered

not considered

1924 Building

Law (kiO.l)

1950. KawaBuml '

a

Selaalc Map

1923 Kanto

4) 192b Specifications for (H-7.9) Design of Roads, Road

Laws, MHA 2nd

3rd

T-12 t

T-8 t

T-6 t

U-600 kg/oi

U-500 kg/m

U-SOO kg/ei

consi- Seismic Coefficient

Method k-0. 15-^0.4 depending on location

and ground cond 1 t Ion

(I&0. 3 advlaed In Tokyo , Yokohama

)

5) 1939 Specifications for Design of Steel High- way Bridges, MHA 2nd

T-13 t

T-9 t

U-500 kg/m

U-400 kg/m

dered

Seismic Coefficient

Method

kh-0.2, kv-0.1

1950. New

Bui ldlng Law (k-0.2, Hil6m

1946 Nankal

( M-8.1)

6) 1956

(and 1964) Revision of Specifications for Design of Steel HIGHWAY Bridges.

JRA

2nd

T-20 t

T-14 t

L-20( 5t/m)

1-14(3. 5t/m)

Seismic Coefficient Method

k-0. 1*0. 35 depending on location

and ground conditions

Increase 0.01

per an increase

of 4m above

16(B))

1948. Fukui

(M-7. 3)

Tokachloki (M-8.2)

7) 1964

to 1971 Specif lcat ions for Design of Substructures

of Highway Bridges. JRA

Sane as 6) same k as (6) Detailed calculation

1964. HighRlse Building.

(Hl45«) h-(0.18M>.36)

1964. Nllgata

(M-7.5)

8) 1966*

(and 1970) Design Standards. JHPC Sane as 6) Design Code for HighRise Bridges, (Increase

in k with the height of

/T(T:perlod)

( 0 1

1

'- 1 - . 2)

1965. Report

9) 1967* Specifications for Earthquake Resistant Dealgn of Honahu-Shlkoku

Bridges, JSCE

Special Live Bridges are Loads and Impact Loads

Considered

for Suspension Modified Selaalc Co- efficient Method

(Basic Coef. k-0.2)

of JSCE on Civil ERD of Engineering

Structures

10) 1967* Specif lcat Ions for Earthquake- Re s 1st ant Dealgn of Metropolitan

Expressway Public Corporation, HEPC

Seismic Coef. Method k-0.2M).3

1968. Report

of JSCE on EarthquakeResistant Design

of Substructures

of Railway Bridges

1966-

okl (M-7 .9)

11) 1968* Aselsmlc Design Standards. HEPCSane as 6) Seismic Coef . Method

k-0.2"U).28

12) 1971 Spec lflcatlons for Earthquake-ReBlstant

Design of Highway

Bridges, JRA

Seismic Coef. Method

k-0. 1-^0.24 (Rigid) Modified SCH.

k-0. 05^0. 3 (Flexible) 1977. A Proposal

for Earthquake

Reslatant Design Methods (HOC)

1979- New Specifi- cations for Esrth- quake-Reslstent

Dealgn of Railway Facilities

1978.

13) 1980 Part V Earthquake

Resistant Design. Specifications for Highway Bridges

(New Specifications)

,

JRA

aue as 6) Seismic Coef. Method

k-0. W). 24 (Rigid) Modified SCH.

k-O.OS-vO. 3 (Flexible) Earthquake Response

Analysis (Very Flexible Bridges)

Mlyaglken- Oki(M-7.4)

* Specifications from 8) to 11) are for highway bridges related to special Large projects.

355
Table 2 Contents of Part V Earthquake Resistant Design, Specifications

for Highway Bridges [Japan Road Association, 1980]

Contents

Same As the Previous

Spec. (1971)

Modified Newly

Slightly Completely Introduced

Chap te r 1 Ge ne r a 1 1.1 Scope O

1.2 Definitions of Terms 0

Chapter 2 Basic Principles for Earthquake

Resistant Design O

Chapter 3 Loads and Conditions in Earthquake Resistant Design

3.1 General o

3.2 Seismic Effects o

3.3 Seismic Forces Transmitted

from Superstructures to Substructures O

3. A Earth Pressures during

Eart hquakes o

3.5 Hydrodynamic Pressures

during Earthquakes o

3.6 Classification of Ground Conditions in Earthquake

Resistant Design o

3.7 Soil Layers Whose Bearing

Capacities are Reduced in Earthquake Resistant Design

O

3. 8 Ground Surface Assumed in Earthquake Resistant Design O

Chapter 4 Design Seismic Coefficient

4.1 General o

4.2 Design Seismic Coefficient

in the Seismic Coefficient

Method o

4.3 Factors for Modifying the Standard Horizontal Design

Seismic Coefficient o

4.4 Design Seismic Coefficient

in the Modified Seismic Coefficient Method Con- sidering Structural

Response

o

4.5 Earthquake Motions in the Earthquake Response Analyses o

4.6 Design Seismic Coefficient Accounting of Ductilities o

Chapter 5 General Provisions for Design of Structural Details

5.1 General o

5.2 Devices for Preventing Superstructures from Falling o

5.3 Vertical Seismic Forces for Design of Connections between Superstructures and Sub- structures o

5.4 Methods for Transmitting

Seismic Forces at Connections

between Superstructures

and Substructures o

5.5 Devices Expected for Decreasing Seismic Forces o

Appendices

I. References on Liquefaction o

II. Examples of Classification of Ground Conditions o

III. Example of Estimation of Design Horizontal Seismic Coefficient o

IV. References on Earthquake Motions

in the Earthquake Response Analyses o

V. Example of Calculating Natural

Periods Utilized in the Modified

Seismic Coefficient Method o

VI. Example of Calculating Ductilities o

VII. Major Related Provisions Concerning Earthquake Resistant Design o

VIII. Detailed Practices o

IX. References o

356
Table 3 Seismic Zone Factor for Highway Bridges

Zone 1 ) Value of vj

A 1.00

B 0.85

C 0.70

Note: 1) Zones A, B and C are illustrated in Fig. 1.

Table 4 Classification of Ground Conditions and Value of v

Group Characteristic Value Tg (second) Value of \>2

1 T

g

< 0.2 0.9

2 0.2 ^ T

g

< 0.4 1.0

3 0.4 ^ T

g

< 0.6 1.1

4 0.6 £ T

g 1.2

Table 5 Importance Factor V3 for General Highway Bridges

Group Definitions Value of V3

1

Bridges on expressway (limited-access highways),

general national highways and principal

prefectural highways.

Important Bridges on general prefectural highways and municipal highways.

1.0

2 Other than the above 0.8

Note: The value of V3 may be increased up to 1.10 for special cases

in Group 1.

357
Table 6 Fl~De Relation

FL Depth, Z(m) Reduction

Factor, Dg

FL <0 .

6

Z i 10 0

10 < Z £ 20 1/3

0 6<Fl^0 8

Z S 10 1/3

10 < Z < 20 2/3

0.8<FL<1.0

Z ^ 10 2/3

10 < Z 1 20 1

1.0<FL 1

Table 7 Magnification Factor 8 for Modified Seismic Coefficient Method

G. C. B-value

Group 1

B = 2T 0.5 ^ T g 0.625

B = 1.25

0.625 1.12

B = 1.40/T

1.12 £ T 1 2.8

B = 0.50

T i 2.8

Group 2

B = 2T 0.5 1 T| 0.625

B = 1.25

0.625 * T i 1.4

B = 1.75/T

1.4 1 T < 3.5

B = 0.50

T I 3.5

Group 3

B = 2T 0.5 = T 1 0.625

B = 1.25

0.625 s T g 1.68

B = 2.10/T

1.68 < T i 4.2

B = 0.50

T i 4.2

Group 4

B = 2T 0.5 i T < 0.625

B = 1.25

0.625 1 T £ 2.0

B = 2.50/T

2.0 S T ^ 5.0

B = 0.50

T 2 5.0

Table 8 Maximum Ductilities from the Analyses of RC Bridge Piers

Section Maximum Ductility Number of

Piers Examined

Circle-Shaped Column 6.4 % 8.1 6 Specimens

Hollowed-Circle Shaped Column 5.8 ^ 6.8 6 Specimens

2-Rectangular Column

Longitudinal 5.6 10.5 6 Specimens

Transverse 5.7 'v 8.6 4 Specimens

Oval-Shaped Column Longitudinal 5.3 ^ 7.3 3 Specimens

358
359
o

Pt o

o

Thickness of Diluvial Deposit (m)

10 20 30 k0 50

10

15

20

25 --a

30

35

U0

Group \/' / Group 2 /

o /

/

/

Group 3

Group k

60

a Group 1 (Ts < 0.25 sec.

)

o Group 2 (0.25 i Ts < 0.5 sec.

)

• Group 3 (0.5 = Ts < 0.75 sec.

)

A Group h (0.75 * Ts)

Fig. 2 Classification of Ground Conditions in Terms of Thicknesses of

Alluvial and Diluvial Soil Layers

360
Thickness of Diluvial Deposit (m) _ 0 20

o

p.uo

0)a

5

10

15

20

25

30

Group

Group 2

Group 3

Previous

Specifications

New

Specifications

Group k

Fig. 3 Comparison of Ground. Conditions Provided

in Previous and New Specifications

N-Value by the Standard Penetration Test

' ' I I I I L_ I L__L

Fig. A Ri-Value (First Term of Resistance R)

361
_ 0.19

-0.05 1

1 i_

r

i i i 1 1 1

1

i i

| 0.6

j i i_i " i, i 01 0.02 0.05 0.1 0.2 0.5 1.0 2.0 5.C

Mean Particle Diameter, D^g-Value on the Grain Size

Accumulation Curve

Fig. 5 R2~Value (Second Term of Resistance R)

2.00

1.5o

1.25

1 .00

0.70

0.5C

0. 30

Group on Ground

Conditions

Y

Grou^ 1

0.5 0.625 1.0 2.0 3.0 U.O 5-0

Natural Periods (T) (sec)

Fig. 6 Magnification Factor for the Modified

Seismic Coefficient Method (New Specifications)

36Z
Fig. 7 Response Spectrum Curves B in case of 7.5 = M = 7.9

7.5UU7.9

20Jmio<60*">

Group 1 of G C. Group 2 Of G c. Group 3 Of G c. Group It of G c.

0.1 0.5 1.0 atural Period (sec.)

ia) 20)01 & Spleentral Distance a < 60tm

7.5«Mi7.9 60kmis<i20l"

0 1

® Group 1 of G C. Group 2 of 0 C. Group 3 of G c. Croup li of G c.

_1_ 0.5 1-0

Netural Periods (sec.)

(b) 60 tan g Epicentre! Distance

7.5SM47.9

1 J0kmia<200ta]

• Group 1 of G C. o Group 2 of G : Group 3 of G c

& Group It of G c

0.5 1-0

Natural Period (»ec.l

120km i Epicentral Pittance & < 200kn

363
Fig. 8 Relationship Between Maximum Horizontal Acceleration (Amax) and

Epicentral Distance (A) Obtained by Statistical Analyses of

Strong-Motion Acceleration Records

364
SINGLE-STORY RESIDENTIAL MASONRY CONSTRUCTION IN UNIFORM

BUILDING CODE SEISMIC ZONE - TENTATIVE RECOMMENDATIONS

by

Ronald J. Morony

ABSTRACT

The U.S. Department of Housing and Urban Development's (HUD) Minimum Property Standards

(MPS) currently require that all masonry construction in the Uniform Building Code (UBC)

Seismic Zone 2 have partial reinforcement.

The housing industry objected that there was not a good basis for this requirement and

it was needlessly adding cost. HUD therefore contracted with the University of California

at Berkeley to establish the behavior of typical single story dwellings under seismic loads

using full scale specimens. This paper is a summary of the four shaking table tests and

timber connection tests previously reported. The form and substance of the tentative

results which were prepared for HUD by the Applied Technology Council of Berkeley are given.

A future fifth shaking test is also discussed.

Upon completion of the contract in middle 1981, a separate report will be presented to

the UJNR Panel.

365
INTRODUCTION

In 1979, Mr. G. Robert Fuller of HUD presented the testing program on single-story

masonry houses in Seismic Zone 2. This report will give a brief summary of that testing

program and then discuss the tentative conclusions and the method of using them. Historically, the research project was caused by complaints from builders in the Phoenix, Arizona

area. In order for houses to be accepted under any of HUD's many programs the construction

must conform to HUD's Minimum Property Standards [1] (MPS). Seismic design requirements are

specified by referring to the Uniform Building Code (UBC) [2]. So in 1973 when UBC changed

their seismic risk map and Phoenix was now in Zone 2 instead of Zone 1 there was a conflict

between the local building code and the MPS. To help clarify the problem a Local Acceptable

Standard (LAS) was issued in August 1974. This LAS did not permit unreinforced masonry and

masonry walls of single-story residences in Seismic Zone 2. The LAS would require at least

the following amounts of vertical reinforcement:

(1) Number 4 bars at all building corners, at wall ends, at all door and window

openings, and in wall sections 2'-0" (0.6 m) wide or less. Openings larger than 12'-0"

(3.7 m) in width require special analysis.

(2) Number 4 bars sp ~ed at not over 12' -0" (3.7 m) on center.

(3) All reinforcement to be matched with dowels embedded in foundation walls or

slabs

.

Local builders questioned the above requirements which are considered to be "partial

reinforcement." The construction contractor's contention was that this reinforcement was

unnecessary and therefore was increasing the construction cost needlessly. The Department

undertook the following research program in order to experimentally determine the actual

minimum requirements.

RESEARCH PROGRAM

There are three parties to the research program. The first is HUD who is sponsoring

the program. Second is the Earthquake Engineering Research Center (EERC) who has performed

the physical testing and is responsible for recommendations to the specific reinforcing

needed. The third party is the Applied Technology Council (ATC). ATC has two tasks, the

366
first is to establish a four man Advisory Panel to assist in monitoring the testing program

and to advise on the preparation of the Guidelines.

The Guidelines will specify and illustrate the method of satisfying the requirements.

It is anticipated that the Guidelines will be referenced in the MPS or the Manual of

Acceptable Practices (MAP) as one acceptable method of satisfying the Seismic requirements.

ATC is also responsible for developing the Guidelines.

The Advisory Panel consists of three consulting structural engineers from three

different geographical areas all of which is Seismic Zone 2 and a representative of the

home building industry in Arizona. [3] The Guidelines are being written by subcontractor

of ATC, Benson and Gerdin Consulting Engineers.

Both the Earthquake Engineering Research Center and the Applied Technology Council are

under contract to HUD.

The EERC ran a series of four tests on their shaking table, Figure 1 [4] is typical

configurations of the specimens used on the shaking table. Note that the shaking table is

capable of one horizontal motion only. It does have the capability of a vertical motion in

conjunction with the horizontal motion.

The single horizontal motion resulted in walls that are either in the plane of the

motion (parallel to the motion) or walls that are out of the plane of the motion (perpendicular to the motion). The walls were 16'-0" (4.9 m) long and 8'-8" (2.6 m) high. The

walls were connected at the top by a timber truss roof diaphram with trusses arranged sometimes parallel and sometimes perpendicular to the horizontal motion of the shaking table.

Concrete slabs were bolted to the roof to account for the reduction in mass from the scaled

down floor plan. Additionally nineteen joint connections were tested.

After three sets of shaking table tests it was determined that the house would not

collapse' under the simulated earthquakes. Therefore, the researchers requested that HUD

establish what constituted structural failure since it was a policy decision. Structural

failure was defined as a permanent displacement of 1/4" (5.8 mm) or greater. This limit was

chosen because it was considered that smaller displacements could be economically repaired

whereas the cost of repair of larger displacements would be relatively expensive. All of

the EERC's recommendations are based upon this failure criterion.

367
EERC TENTATIVE RECOMMENDATIONS

The problem of making recommendations become more complex because during the course of

testing a new seismic risk map was published. This is contained in the "Tentative Provisions for the Development of Seismic Regulations for Buildings" [5] (ATC). It was decided

to use the accelerations from ATC-3 and apply this criteria to the areas shown in the UBC

1973 Seismic Zone 2. This overlay is shown in Figure 2. The ATC-3 document speaks to

Effective Peak Accelerations (EPA) which are related to the seismic spectrum. The shaking

table motion having a peak acceleration of 1 g (acceleration of gravity) is assumed to have

an EPA of 0.82 g. The maximum EPA of 0.2 g indicated by the ATC-3 for UBC Zone 2 is represented by a peak shaking table acceleration of 0.24 g [6]. Because of the large range of

accelerations in Seismic Zone 2, Seismic Zone 2 was subdivided into Zone 2A with a range

of EPA = 0.1 g or less and Zone 2B with a range of EPA = 0.2 g. The zones are shown on

Figure 3.

The recommendation for Zone 2A is very simple. No reinforcement in single-story

residential buildings of standard clay brick or concrete block construction is required for

earthquake resistance [6].

Zone 2B will require some partial reinforcement in masonry walls. The requirements

will be specified in the Guidelines. The Guidelines are meant to be used by home builders

and therefore are a "how-to-do-it" manual. The use of the manual will be restricted to

houses that are "regular" in plan and meet the following requirements:

(1) The area under roof must not be greater than 3,400 square feet (315 m)

.

(2) The total weight of roofing, ceiling, insulation and other materials

supported directly by the roof trusses or rafters must not exceed 20 pounds per square foot

(100 kg/m2 ).

(3) All masonry walls must be constructed with a concrete or masonry stem wall on a

concrete footing.

(4) The unsupported height of any wall above the floor must be no greater than 8'-4"

(2.5 m) and the height of any parapet above the roof must be no greater than 3'-0" (0.9 m)

.

(5) The roof must be plywood and mailed to the wood framing members (trusses or

rafters)

.

368
(6) Roof trusses bearing on the top of the unit masonry unit must not have a

clean span greater than 40'-0" (12.2 m)

.

(7) The roof rafters, joists, and trusses which are supported from a ledger bolted

to the face of the masonry wall must not have a clean span greater than 20' -0" (6.1 m)

.

If the house does not fall within these limitations, the Guideline is not applicable

and the house should be reviewed by a registered structural engineer for earthquake

resistance. [7]

The Guidelines would illustrate the response of houses to earthquakes such as Figure

4. It will show how to resist these forces such as Figure 5. It will also illustrate how

to make connections between the timber members and the masonry such as is shown in Figure

6. This general approach is taken so that the homebuilder may readily use the Guidelines.

For clarity the Guidelines will be printed in two colors so that the non-conventional

construction will be readily apparent. The requirements and location of shear panels is

shown in Figure 7

.

The Guidelines have not been finalized because there are some critical issues that

could not be resolved by the Project Advisory Panel. Because of the physical limitations

of the shaking table there was data on walls in-plane and walls out-of-plane . In a real

earthquake all walls will be both in-plane and out-of-plane at the same time. Another

problem was that using accepted analytical methods the number 3 bars should have broken

but they did not. This fact raised a concern that the entire research program could be

seriously questioned on this account. And finally, a concensus could not be reached as to

the required length of shear panel and the effects of tension on placement of shear panels.

It appeared that the vast majority of these issues could be resolved if an additional test

were performed. This test would subject the walls to a combination of in-plane and

out-of-plane forces simultaneously.

ADDITIONAL RESEARCH

The testing program has significantly increased the state-of-the-art and has provided

for the first time, quantitative as well as qualitative data on the response of typical

single-story masonry houses to earthquakes. It is the opinion of the researchers, the

Project Advisory Panel and the HUD representatives that most aspects of the behavior of

real houses during actual earthquakes have been adequately simulated during the research

369
program. To quote from the final report by the EERC "... this report has emphasized that

one important feature of the real earthquake situation has not been included in these

tests: biaxial horizontal excitation. No shaking table in the United States has the

capacity of testing structures of the size of these house models with biaxial horizontal

motions, so it is not possible this biaxial effect with complete similitude ... if a test

structure ... were positioned on the shaking table with one of its axes rotated 30 degrees

relative to the excitation axis, then the single horizontal earthquake component would

induce both in-plane and out-of-plane forces in all walls." [8]

The possibility of the additional test was discussed with the Project Advisory Panel

in February 1980. The panel was of the opinion that this additional test could resolve the

problem of combining the in-plane and out-of-plane effects, the apparent over stressing of

the number 3 bar and the attendant determination of the minimum length of shear panels.

The Advisory Panel suggested that a 40 degree orientation might be more advantageous since

a cursory review of accellometer readings indicated that the minor axis accelerations were

approximately 80 percent of the major axis accelerations. It was also suggested that the

reinforcement be instrumented with strain gages to determine actual stresses in the

reinforcement

.

CONCLUSIONS

It is the explicit goal of the Department to set standards at the lowest reasonable

level. Therefore every effort is being made to either sponsor or obtain funds to perform

the fifth test described. It is anticipated that this test will be performed in the last

part of 1980 or in early 1981. Copies of the first two volumes are currently available

from the National Technical Information Service (NTIS) Springfield, Virginia 22161 USA.

The third volume "Summary, Conclusions and Recommendations" should be available through

NTIS shortly.

Films of the first four tests and the fifth shaking table will be edited into a

concise presentation and should be available for future UJNR Panel meetings.

The Guidelines will be published as a two color document by HUD's Office of Policy

Development and Research. It is anticipated that the Guidelines will be referenced in

either HUD's Minimum Property Standards (MPS) or in the Manual of Acceptable Practice

(MAP)

.

370
REFERENCES

U.S. Department of Housing and Urban Development, "Minimum Property Standards, Oneand Two-Family Dwellings," HUD 4900.1, 1973 Edition, Washington, D.C.

International Conference of Building Officials (ICBO), Uniform Building Code, Vol. L, 1973 Edition, Whittier, CA.

The three consulting engineers are 0. Clarke Mann, Memphis, Tennessee, James P. Kesler

Pasadena, California, and Joseph A. Gervasio, Phoenix, Arizona. The home building is

Lenn Pritchard, a masonry contractor.

Ray W. Clough, Ronald L. Mayes and Polat Gulkan, "Shaking Table Study of Single-Story

Masonry Houses; Volume 3: Summary, Conclusions and Recommendations," Report No. UCB/EERC 79/25 September 1979.

Applied Technology Council Publication ATC 3-06, "Tentative Provisions for the

Development of Seismic Regulations for Buildings," (NSF Publication 78-8, NBS Special

Publication 510) U.S. Government Printing Office, June 1978.

Ray W. Clough, Ronald L. Mayes and Polat Gulkan, "Shaking Table Study of Single-Story

Masonry Houses, Volume 3, Summary, Conclusions and Recommendations," Report No. UCB/EERC 79125, September 1979.

Benson and Gerdin, Consulting Engineers for the Applied Technology Council,

"Guidelines for One-Story Houses in Seismic Zone 2" (Draft) January 15, 1980

unpublished

.

Ray W. Clough, Ronald L. Mayes and Polat Gulkan, "Shaking Table Study of Single-Story

Masonry Houses; Volume 3: Summary, Conclusions and Recommendations," Report No. UCB/EERC 79125, September 1979.

371
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372
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FIGURE 4

374
FIGURE 5 Earthquake Resistance System

37 5
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376
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Zone 2B- Plan I- Exterior Elevations
EARTHQUAKE DISASTER PREPAREDNESS PROGRAM FOR THE LIMA, PERU METROPOLITAN AREA

by

Masamitsu Ohashi

Makoto Watabe

Yoshio Kumagai

Yoshihisa Hoshino

INTRODUCTION

Disaster prevention in major earthquakes is a most important national project in

earthquake-prone countries. Though some minor differences can be observed in seismological

conditions between Peru and Japan, both countries are quite similar. Based on this fact,

our technical cooperation will be quite fruitful for disaster prevention projects in both

countries

.

In spite of great advances in the technology of predicting major earthquakes,

prediction at present still is not completely dependable. Therefore, the best method to

prevent earthquake hazards is to prepare all possible countermeasures against major earthquakes, namely, disaster prevention projects.

By establishing disaster prevention systems, earthquake hazards will be reduced

drastically even for the great earthquake expected in the future.

Twice, our technical cooperation mission from Japan has been to Peru. The first

mission was sent to Peru in 1979. It consisted of three members, Mr. Masamitsu Ohashi,

the leader of the mission, Mr. T. Matsui, and Mr. Y. Kitani. The mission stayed in Peru

about ten days to investigate the types of technical cooperation that might be most

effective for disaster prevention in the Lima Metropolitan Area.

Based upon the report of the first mission, the second technical mission was sent

to Peru in late 1980. The second mission stayed in Peru about three weeks and was

mainly devoted to the collection of information, meeting with persons concerning disaster

prevention for major earthquakes. These visits benefitted from the courtesy and kindness

of the Civil Defense which played the role of the counterpart organization in Peru.

Names and specialized fields of the members of the second mission, as well as their

schedule in Peru are as follows:

378
(1) Makoto Watabe, Mission Leader, Earthquake engineer, Dr. of Engineering

(2) Yoshio Kumagai Urban plan specialist, Dr. of Engineering

(3) Yoshihisa Hoshino Geologist

(4) Kazuhiko Kawashima Civil Engineer

SCHEDULE

November

22 Arrived at Lima

23 Courtesy call on Minister of Interior and Director of Civil Defense

26 Meeting at Civil Defense

28 Meeting at Geophysical Institute

29 Meeting at Geological Institute

30 Meeting at La Molina University

December

3 Meeting at UNI

4 Meeting at Catholic University

5 Meeting at Ministry of Housing and Construction

6 Evaluation at Civil Defense

7 Visiting Arequipa to meet with the engineer

8 Visiting Cusco, courtesy call on the mayor

10 Preparation of report

11 Lecture to staff of Ministry of Housing and Construction, and lecture to staff of

Civil Defense later in the day

12 Discussion at Catholic University

13 Discussion at UNI

14 Submitting report to the Minister of Interior.

The report contents include the following items: (1) City planning for disaster

prevention, (2) Geological works, (3) Earthquake ground motions, (4) Microzonation,

(5) Civil Engineering structures, (6) Housing and buildings, (7) Civil Defense.

Here, all members of the mission would like to express the hearty gratitude to

Colonel EP Heraclio Fernandez Pendola, Director Superior del Comite National de Defensa

379
Civil for his warm hospitality and cooperation, and to Colonel PIP Guillermo Castillo

Rivadeneyra, Jefe del Comite de Asesoramiento del Comite National de Defensa Civil;

Ingeniero Julio Kuroiwa Horiuchi, Presidente del Comite1 Cientifico de la Secretaria

Ejecutiva de Defensa Civil; Ingeniero Guillermo Chamorro Rodriguez; and all the staff of

the Comite National de Defensa Civil for their cooperation.

CITY PLANNING FOR DISASTER PREVENTION

One aim of this project, Disaster Prevention in Lima Metropolitan Area, is the

formulation of a basic plan for disaster prevention. The first year, 1979, objectives

of this technical cooperation are as follows:

Develop comprehensive planning system for the regional and city plan in Peru and

in the Lima Metropolitan Area,

Provide suggestions for earthquake disaster countermeasures and earthquake

disaster prevention plans in the Lima Metropolitan Area,

Present the contents and schedule for this technical cooperation in the

future

.

For these purposes, we held discussions with members of the National Committee of

Civil Defense, Ministry of Housing and Construction, National University of Engineering,

and others. The suggestions to these groups were as follows.

To the National Committee of Civil Defense

Estimate the degree of danger of a large scale earthquake in the Lima

Metropolitan Area

Establish a public relations system for earthquake disasters

Coordinate and arrange a earthquake disaster prevention plan for each

ministry

Establish a radio communication network system for the ministry

Provide planning provisions for emergency support from abroad

Designate emergency evacuation sites.

To the Ministry of Housing and Construction

Establish a new type of zoning system for building structure control

Initiate model planning for a small plaza in a concentrated block of "adobe'

construction

380
Designate reconstruction sites for post large-scale earthquake activities.

To other ministries

The intensification of the official car check, system

Designate a traffic control zone for post large-scale earthquake activities

Establish a fail safe system for public facilities

Concentration of governmental office buildings

These suggestions are further detailed in Table 1.

Other suggestions are shown in Figures 1 and 2. Regarding a new type of zoning system

for building structure control, we discussed the four following types:

(1) Aspects of important zones

(a) Around emergency facilities

Obligation to build earthquake resistant buildings within 10 or 15

years

° The necessity to subsidize people for reconstruction.

(b) Important zones in relation to urban activities

Control of the building structure in new construction, reconstruction,

and remodeling

(2) Aspects of microzonation

(a) Urban activities and microzonation

0 Control of building structure in new construction, reconstruction,

remodeling, and after a large scale earthquake

(b) Zones estimated to experience huge damage in large scale earthquakes

° Prohibition of new construction after large scale earthquakes

0 Preparation and designation of reconstruction sites

° Prohibition of new construction in designated reconstruction sites

before large scale earthquakes

GEOLOGICAL WORK

The geological data, including the geomorphological and geodesical, are important for

microzonation. We recommended three works on the geology. They are the Landform Classification Map , the Superficial Geological Map , and the Neotectortic (Quaterary Tectonics)

research.

381
Landform Classification Map

From our surveys in Japan, we are able to recognize interrelationships between

earthquake damage patterns and landforms. This derives from the fact that some coincidence

can be seen between landforms and ground conditions, since each landform, either erosional

or depositional , has an individual internal structure which depends on the forming process.

This interrelation can be used in reverse, and geologically high-risk areas can be

disclosed qualitatively.

A Landform Classification Map should be planned using a scale of 1:50,000 or larger.

The scale of 1:25,000 is better.

Superficial Geological Map

In addition to the Landform Classification Map, the Superficial Geological Map is

suggested. Useful information about the superficial ground condition can be obtained by

making a Superficial Geological Map.

Boring data is needed to develop the Superficial Geological Map. Therefore, boring

data in and around Lima should be collected and compiled as soon as possible.

Neotectonics (Quaternary Tectonics)

For microzonation, information about crustal movements of recent time is important.

Recent crustal movement has continued to present time since the early Quaternary period.

Therefore, research on Quaternary Tectonics should be carried out. The crustal movements

of past and present time can be investigated using the three methods: the geological

method, the geomorphological , and the geodesical. The geological method can make clear the

crustal movement before 10^~^ years B.P. The geomorphological method is effective in

learning of crustal movements during 105~6 - \q3 b.P. While, the geodesical method is

indispensable to get data about the crustal movement after 10^ years B.P.

The crustal movement as revealed by the geological method is only the old structure in

general which has not always succeeded to the Quaternary period. Therefore, the geological

method is sometimes not useful to study Quaternary Tectonics. Instead, the geomorphological

method is effective for such purposes.

The geodesical method is fundamental in deciphering present crustal movement. A survey

of trig stations and bench marks will be recommended as a beginning point, not for making

maps but to get data about present crustal movement. However, the geodesical method covers

382
the tendency of crustal movement for only 100 years at most, despite the fact that the

crustal movement sometimes has a periodicity of more than 100 years.

Therefore, the geomorphological method and geodesical method should be coordinated to

develop the recent tectonic situation. The research of geomorphology in general, and the

geodesic survey in and around Lima will be recommended for the long term efforts.

EARTHQUAKE GROUND MOTIONS

(1) According to several accelerograms obtained here in Peru, the characteristics of

earthquake ground motions, such as peak accelerations and shapes, are quite different from

those observed in Japan and the United States. Therefore, an extensive collection of strong

motion accelerograms in Peru is urged, so that Peruvian characteristics of earthquake ground

motions may be established.

(2) In view of very high peak accelerations in Peru, studies of earthquake ground

motions which correlate observed physical values (such as effective peak accelerations) and

those estimated through investigation of damaged structures may be important for the interpretation of observed accelerograms. Also, re-evaluation of peak accelerations of original

accelerogram records is suggested.

(3) In order to collect as many strong motion accelerograms as possible, an

instrumentation plan for strong motion accelerographs on the ground as well as on specific

structures will be necessary: this instrumentation plan should cover a short, intermediate,

and long term scheme with an order of priorities.

(4) Strong motion accelerograph arrays are recommended for instrumentation in such a

place as La Molina, where some heavier damage has been reported in past earthquakes, and

where detailed study on response behavior of soil-deposits due to earthquake excitations

appears necessary.

(5) Positive establishment of systems for faster digitization and analyses of

accelerograms obtained as well as faster supply procedures for these results to concerned

practical researchers and engineers is strongly recommended.

383
MICROZONATION

General

Through past experience with earthquake damage, it is known that there are clear

differences between severely damaged areas and slightly-damaged areas even within a relatively small area. Such evidence is clear in the La Molina district of the Lima Metropolitan Area since the MM intensity in the district has been two to three grades stronger in

the 1904, 1932, 1940, 1966 and 1974 earthquakes. It is considered important to define the

area where damage is expected in the future earthquake using the microzonation technique.

From an earthquake disaster prevention point of view, the following microzonation might be

ef f ective.

(1) Subsurface ground condition

(2) Distribution of SH-wave velocity

(3) Amplification factor of subsurface ground

(4) Liquefaction

(5) Land slide

(6) Slope failure

(7) Water table

(8) Distribution of building construction zone maps

(9) Tsunami

For the above purposes, two approaches are proposed for microzonation, i.e., an

approach based on investigations of past earthquake damage, and the theoretical approach.

Recommended Approach for Microzonation in Lima Metropolitan Area

In order to conduct microzonation in the Lima Metropolitan Area, the theoretical

approach based on the calculations of subsurface ground motions is recommended, and then to

re-check and improve the calculated results from the knowledge obtained by the investigations of past earthquake damage features. The following subjects are recommended for

investigation using the theoretical approach:

(1) Research on the dynamic characteristics of subsoils

Improved test procedures and the results concerning dynamic properties of subsoils

have been developed in recent years. Also recommended is the acquisition of test

384
apparatus and study of the test procedures in order to investigate the following

subjects:

(a) Dynamic shear moduli

(b) Damping characteristics

(c) Dynamic shear strength

(d) Sampling technique to get undisturbed soil specimens

The introduction of the resonant column apparatus and the dynamic triaxial test

unit are suggested to pursue these investigations.

i CIVIL ENGINEERING STRUCTURES

Transporation Facilities

Roads

It is suggested that roads be classified into several groups in accordance with their

functions after the earthquake hazard. The important functions, indispensable for postearthquake activities, might be as follows:

(1) Roads for refuge

(2) Roads for rescue

(3) Roads for fire emergencies

(4) Roads for transportation needed for restoration

(5) Roads for transportation of food and water

For such functions, it is suggested the following points be examined:

(a) Check earthquake resistance of road facilities, including traffic control

equipment

(b) Survey for possible bottleneck points

(c) Reinforcement and retrofitting procedures to upgrade marginal roads

(d) Preparation of duplicate roads for important highways

Bridges

Certain types of bridges are quite susceptible to earthquake damage. Seismic

damage is most commonly caused by foundation failures resulting from excessive ground

deformation and/or loss of stability and bearing capacity of the foundation soils. As a

385
direct result, substructures often tilt, settle, slide, or even overturn. Severe cracking,

or complete failure, does occur.

A survey of present bridge design and construction practice is suggested, paying

special attention to earthquake resistance of foundations and foundation soils.

The establishment of earthquake resistant design methods for bridges, and the

investigation of retrofitting methods is recommended.

Ports and Harbors

A survey of present design and construction practice of ports and harbors is suggested.

The establishment of earthquake resistant design methods for ports and harbors is

recommended

.

Dams and Reservoirs

A survey of present design and construction practice for dams and reservoirs, which

are used for irrigation purposes, generation of electricity, and water supply is suggested.

The establishment of earthquake resistant design methods for dams and reservoirs is

recommended.

River and Coast Dykes

A survey of the present condition of river and coast dykes, especially where floods or

Tsunamis may occur, is suggested.

Underground Structures

Underground structures such as conduits and pipe-lines are widely used for the

supply of electricity, water, petroleum, the distribution of telecommunications, etc. From

past earthquake damage, it is often observed that the damage to underground structures

takes place where the structural or geological conditions change discontinuously

.

A survey of present design and construction practice for underground structures is

suggested. The establishment of earthquake resistant design methods for underground

structures important to post-earthquake activities is recommended.

Tunnels

From past earthquake damage, it is often seen that the interruption of traffic in

tunnels is not caused by damage to the tunnels themselves but by slope-failures at the

386
ends of the tunnels. A check of the present condition of tunnels paying special attention

to slope-stability at the end of tunnels is suggested.

HOUSING AND BUILDINGS

(1) It surprised us to see that many adobe structures have survived with little damage

the various strong motion earthquakes of the past 200 years. In view of this, it is suggested that simple local materials such as adobe or brick blocks be used for low-cost

housing, if such materials are used properly.

(2) Extensive research and development projects on low-cost housing are underway;

from what we gathered in visiting two major Lima universities, the philosophy and orientation of the projects are quite satisfactory. Most possible methodologies for strengthening

adobe walls and their connections have been explored. It is, however, also true that with

some moderate increase in experimental instruments at these facilities, the efficiency and

accuracy of research work might be dramatically improved.

(3) Essential experimental equipment and facilities such as microtremor measurement

instruments, hydraulic actuators, dynamic strain-meters, and shaking tables should be

quickly supplied.

(4) Development of a low-cost house with one very strong room against earthquakes,

research on the connections between walls, and shear connections between adobe walls and

wooden braced frames, may be a useful focus.

(5) Public dissemination of research results on developments mentioned above are

vital to improve the low-cost housing. Various investigations of possibilities such as

tax-reductions, economical structure investigations, land-use advantages, or training

projects for builders and engineers should be studied to encourage people to adopt the

recommendations

.

(6) Retrofitting buildings and housing against possible strong motion earthquakes is

urged, especially for governmental office buildings. For this purpose, evaluation criteria

of the aseismic capacities of buildings and reinforcing procedures for these buildings

should be investigated. It is also important to educate engineers in the evaluation of the

degree of diminished safety in damaged buildings after earthquakes.

(7) In order to coordinate practical and precise earthquake resistant regulations for

buildings, the following back-up research work is needed:

387
(a) Micro-tremor measurement of natural periods and damping characteristics of

building structures

(b) Behavior of infill-wall structures

(c) Effects of shape factors of structures such as the torsional and

discontinuity effects

(d) Research on the behavior of non-structural elements

(e) Research on the ductility of various structural types

CIVIL DEFENSE

In Japan, we have no permanent and independent organization for disaster prevention.

The National Land Agency does include a Disaster Prevention Policy Planning Division, and

the division manages the Central Disaster Prevention Council. The council is established

by law, the Disaster Countermeasure Basic Act. Based on the act, local governments

establish local disaster prevention councils. When a big disaster, such as storm, flood,

or severe earthquake occurs in a region, the national and local governments cooperatively

establish disaster prevention headquarters based on the disaster prevention plan.

In Peru on the other hand, there is a permanent and independent organization for

disaster prevention. Few countries in the world have an organization such as Defensa Civil,

so there is good reason to be proud of it. We do recommend several items to reinforcing

its function.

The first point relates to budget and scope. Individual items for disaster

prevention, such as adobe construction, observations of strong motion etc., are pursued in ,

the universities and institutes. These are of high level importance, but there are few

comprehensive projects for disaster prevention such as the estimation of dangerous degrees,

damage estimations, and designation of evacuation sites, etc. So, the Civil Defense should

take a leading role in such comprehensive work in cooperation with the universities and

institutes. For this purpose, we recommend budget increases for such disaster prevention

activities and to increase the number of full-time technical staff members.

The second point relates to earthquake preparedness in the other Civil Defense

Headquarters itself. Defensa Civil has an exclusive building which will be effective in

countermeasures for a severe earthquake. however, if the building itself is damaged by

such an earthquake, the abilities of the Civil Defense, to lead will suffer. So, we

388
recommend the reinforcement of the building structure, securing antennas and other important

instruments, such as radio apparatus, attaching heavy furniture — such as bookshelves, and

steel lockers — to walls to prevent their overturning and destruction.

CONCLUDING REMARKS

Comments and recommendations on basic concepts of disaster prevention against major

earthquake hazards in the Lima Metropolitan Area, are introduced in previous section. As

indicated in section summaries, research needs, tasks, projects and works are classified in

short term (.approximately one year), intermediate (approximately three years), and long

term (approximately more than three years).

Although all of these items are essential for a project of disaster prevention in the

Lima Metropolitan Area, the following items should be strongly emphasized for the short

term;

(1) Designation of model districts, such as Callao City and the Lima District, for

degree of danger estimates

(2) Division of model districts into several zones according to state-of-art

microzonation procedures

(3) Collection of urban statistical data in each zone

(4) Develop a Landform Classification Map for microzonation in and around the Lima

Metropolitan Area

(5) Collect boring data in specific districts such as Callao, La Molina and Lima

(b) Establish a shared publication system of accelerograms and analytical

results for research and practical engineers

(7) Develop installation plans for accelerographs on the ground and in structures

with a priority order for engineering purposes

(b) Survey subsurface ground conditions, including the measurement of Sh-wave

velocity in the specific districts such as Callao, La Molina and Lima

(9; Initiate research on the dynamic properties of subsoils

(10) Begin organization of post-earthquake transportation systems and survey for

bottleneck points in the total transportation system

(11) Survey the present design and construction practices in the civil engineering of

structures

389
(12) Establish a research center for earthquake disaster prevention in Peru.

(.13) Establish an evaluation system for aseisinic capacity of k.C, masonry, and adobe

structures

(14) Retain at least three full-time engineers in the Comite National of the civil

Defense

In order to implement the above, a personnel exchange not only from Japan to Peru, but from

Peru to Japan in as early a stage as possible is strongly emphasized.

390
Table 1. CITY PLANNING RESEARCH, TASKS AND PROJECTS FOR DISASTER PREVENTION

Short Term Intermediate Term Long Term

1. Study on designation of evacuation sites

2. Study on designation of model district and division into zones to estimate degree of jsearch danger

3. Study to estimate degree of danger Continue 3 Continue 3

4. Study of public relation system for earthquake disaster

Continue 4

5. Study of new zoning system for control of building structures

b. Designate evacuation sites in Lima

Metropolitan Area

Planning in concentration of governmental office buildings

7.

8.

Establish radio communication network system

among ministries

Intensification of offical car check system Continue 8

9. Designation of traffic control zones after

Tasks large scale earthquakes

10. Collecting statistical data in each model

district zone Collection of statistical data

in the whole of Lima Metropolitan

Area

11. Estimation of degree of danger in model

district

Estimation of degree of danger in the whole of Lima Metropolitan

Area

12. Designation of new building structure control

zone Continue 12

13. Model planning of a small plaza in an "adobe" - construction-concentrated zone Carrying out 13 Concentration of governmental office

buildings

rojects 14. Establish fail-safe system for public

facilities

Continue 14

or 15. Preparation of reconstruction sites Designation of reconstruction sites

|

Continue 15 lb. Planning provisions for emergency support Arrangement of lb Works from abroad

17. Establish public relation system for earthquake disaster

Continue 17 Continue 17

18. Formulating earthquake disaster prevention

plans in each ministry

Continue 18

391
EARTHQUAKE GROUND MOTIONS

Short Term Intermediate Term Long Term

Research

Data processing

procedures for

accelerograms

Effective peak acc. E.P.A. concept

Correlation study,

observed and estimated

kisk analysis in view

of E.P.A. and spectra

The same as item at left

Tasks

Publication of

strong motion

accelerograms

(digitized and

spectra obtained

in Peru

Instrumentation

plan with priority, ground, and

structure

Strong motion acc. arrays observation

The same as item at left

Projects

or

Work

Establishment of

faster digitization, analyses,

and publication

systems

Installation of

strong motion

accelerographs

Strong motion acc. array installations

in La holina

The same as item at left

Collection of strong

motion accelerograms

The same as item at left

392
RESEARCH, TASKS, AND PROJECTS FOR MICROZONATION

Short Term Intermediate Term Long Term

Research

Investigation of

earthquake damage,

liquefaction, land

slide, slope failure, Tsunami, etc.

| Calculation of subsoil

| deposits for strain

| dependency of shear

| moduli and damping

| characteristics

| Effects of ground

| condition

| Classification of

| ground condition

Effects of topographical condition

Evaluation of liquefaction potential

Evaluation of land

slide potential

Collection of

boring data

Sampling of undisturbed soil

Measure of SH-wave

velocity

Classification of

ground condition

Tasks

Distribution of

building

construction maps

Collection of

literature and

information con- cerning earthquake

damages

Preparation of

questionnaires and

forms on damage

ground motion

| Same as left

Projects

| Experimental work

| concerning dynamic

| shear moduli and

| damping character-

| istics of subsoil

| using resonance

| column apparatus

Collection of uata

concerning dynamic

strength of subsoils

using dynamic triaxial test apparatus

393
RESEARCH , TASKS, AND PROJECTS FOR CIVIL ENGINEERING STRUCTURES

Short Term Intermediate Term Long Term

Research

Survey of earthquake resistant

design methods

Seismic response

characteristics of

facilities such as dams, bridges, high

embankments , retaining

walls, port, harbor,

and underground

structures

Research on earthquake resistant

design procedures

Research on retrofitting methods

Tasks

Definition and | Determination of

selection of | design forces

important

transportation | roads

Checking facilities | Collection strong

and present design | motion earthquake

and construction | records for each

practice | structure

Listing bottle neck | Research foundation

points on roads | structure design

vital to total | procedures

transportation

system

Projects

1 1

1 1

1 1

Establish earthquake

resistant design

methods

Establish retrofitting methods

394
HOUSING AND BUILDINGS

Short Term Intermediate Term Long Term

IV L O V tl L L 1 1

Microtremor

analyses

Shape factor

rpcparph on 1 nu

cost housing

Evaluation method

for aseismic capacities of adobe,

masonry and k.C.

Methodology for

masonry and k.C.

structure repair work

T1111 n f i1111 1 —T.7o Wall1 1 ct"T"iir o

, t~i1T'fi L L ULUULC

Non-structural

element equipment

Ductility of R.C. and macnnrv Qt~T"iir, t_ LUCIO \JLL L iiT"£*c y J LLULLULCO

Communication plans

for developed lowcost housing

Development of lowcost prefabricated

house

Building code

development

Tasks Aseismic capacity

evaluation of important buildings

Collecting building

code research results

Proj ect

or

Work

Plan for Peru

Earthquake Engineering Research

Center

Retrofitting important

buildings

Establish Earthquake

Research Center with

basic equipment

Continue at left

Continue at left

395
Fig. 1 Flow Chart Estimate of Danger Degrees in the Lima Metropolitan Area

Population and

number of households'

Microzonation

Building structure

and Building use

Distribution of

public facilities

Designation of model districts

1 r Division to small zones

in model districts

5 9 Collection of statistical

data in each zone

r Estimation of degree of

danger in model districts

1 Division to small zones

in the whole of Lima

Metropolitan Area

Example

Lima District

Callao City

• Trend of population

• Composition of population and

households

• Dangerous

facilities

• Disaster Prevention

facilities

Collection of statistical

data in each zone in the whole of Lima Metropolitan

Area

r Estimation of degree of danger in the whole of Lima

Metropolitan Area

r Execution of map of degree

of danger in the whole of

Lima Metropolitan Area

T

Formulation of earthquake

disaster prevention plan in

Lima Metropolitan Area

396
Fig. 2 Technical Cooperation Schedule on Disaster Prevention City Planning

1978, Pre-survey Committee

* Dispatch of experts

• Acceptance of counterparts

1979, Technical Cooperation

• Necessary of estimation of degree

of danger

* Data for the estimation

1980, Technical Cooperation

• Method of estimation of degree

of danger

• Estimation of degree of danger

in model district

1981, Technical Cooperation

• Estimation of degree of danger

in Lima Metropolitan Area

* Disaster prevention basic plan

1982, Technical Cooperation

• Formulation of disaster preven- tion basic plan in Lima

Metropolitan Area

Request of Technical

Cooperation

Preparation of Technical

Cooperation

Designation of model district

Division to small zones in

model district

Collection of statistical data

Division to small zone in the

whole of Lima Metropolitan Area

J|* Collection of statistical data

in each zone

Execution of map of degree of

danger in Lima Metropolitan

Area Collection of data in relation

to disaster prevention plan

Japanese Team Peruvian Team

397
FUNDAMENTAL FACTORS IN OPTIMIZING EARTHQUAKE

DISASTER MITIGATION INVESTMENTS

by

Yoshijiro Sakagami

Eiichi Kuribayashi

Osamu Ueda

Tadayuki Tazaki

INTRODUCTION

Earthquake disaster mitigation investment become more reasonable by comparing invested

resources to expected effects. Unlimited investment is not possible, for one reason,

because severe earthquakes occur so infrequently that the new resources do not yield sufficient expected effect. Moreover, there are limited resources to invest.

Earthquakes are unusual phenomena, so disaster mitigation investment plans should not

be alien to the usual regional plans. Inconvenience and inefficiency may not always be

prime factors in considering disaster mitigation investments. This paper discusses a number of the fundamental factors in choosing earthquake disaster mitigation investments.

For the purposes above, it is necessary to estimate damage amounts and probability.

Damage is classified as follows:

(.1) Physical loss

(.2) Economic loss

(3) Social damage

Social damage is difficult to estimate because it is complicated by political,

sociological, and psychological implications. Therefore, this paper deals mostly with

physical and economic losses, and decision making procedures.

ESTIMATION OF PHYSICAL LOSS

Property Loss

Property loss is defined as the reduced value of property due to earthquakes.

Estimating methods of property loss are classified in two ways: First, the loss is correlated with earthquake dimensions and ground conditions by using data from past, damaging

earthquakes to estimate the loss from future earthquakes by using the correlation. Second,

398
the mechanical fracture of a structure and the ultimate seismic loading is defined, and the

probability of the occurrence of an ultimate loading is estimated. The index of property

loss is defined as the damage ratio to an individual facility or as the loss ratio valued

in money. The damage ratio of an individual facility is used to assess the effectiveness

of strengthening and retrofitting the vulnerable aspects. However, this is not adequate to

assess regional disaster resistivity, because it does not represent the total damage of the

region. Neither can it be compared with that of other individual facilities. To establish

regional resistivity, the loss ratio which is the ratio of the loss to existing assets,

both valued in money, is more useful. [1] This ratio can be used to compare the resistivity of different facilities, to calculate the ratio for total facilities, and for application in evaluating economic loss.

In evaluating the loss ratio of past earthquakes, property loss, which has two

definitions, should be used. One is the replacement cost. The second is the value in

current prices which takes depreciation into consideration. The former is mostly applied

to public facilities, and the latter mostly to private property. Both methods have not yet

been standardized in a official manner, and the data might contain a certain range of

error

.

Functional Loss

Lifeline facilities — such as road, railroad, electric power, telecommunication,

water supply, sewerage, and natural gas supply — are infrastructures which supply goods,

energy, and information to the society. If they are damaged, the functional loss is more

important than the property loss. Functional loss delays the social and economic activities which are supported by lifeline facilities. One delay causes another delay in growing

repetition.

The relationships between recovery time of lifelines as an index of the functional

loss and severity of a disaster have been studied in several natural disasters. 1 2 , 3

J

Recovery time is determined not only by the property loss, but it also is affected by the

difficulty of recovery which depends on topographical and geological conditions, equipment,

material, and labor forces for recovery. High density of lifeline activities due to

urbanization may further slow recovery.

399
Tables 1 and 2 summarize the recovery time for electric power and natural gas supply

facilities in past Japanese earthquakes. Taking as examples the Niigata and Miyagiken-oki

earthquakes, the magnitude and the size of the quake-hit cities of Niigata and Sendai —

population as of the year of the earthquake was 360, UUO and 620, UUO respectively — of the

two earthquakes are roughly equivalent. Epicentral distances to the two cities were 50 km

and 110 km respectively, kecovery time in Sendai is much shorter than in Niigata. The

reason is that redundancy in electric lines and the inter-gas company emergency cooperation

system had been improved after the Niigata earthquake.

ECONOMIC LOSS

Property and functional loss affect the regional economy by causing damage to the

machinery of industry and reducing productivity. As an example of the biggest consequence,

Kato refers to the aftermath of the Kanto earthquake when a recession was accelerated and

the government faced difficulty in finance and international trade. [4] However, in

recent earthquakes we have not had a national-scale consequence. In many cases the

economic consequence was hard to recognize statistically even at the prefectural level.

One reason is that the recent earthquakes were not as big when compared to the Kanto earthquake. Another reason is that an earthquake has a positive influence in the economy as

well as the negative one. An earthquake generates additional demand of material, equipment, and labor for recovery and reconstruction. It a surplus of labor force and equipment exists, the impact of an earthquake may absorb them.

In the Alaska earthquake of 1964 it is reported that the earthquake damage generated

new demand for reconstruction which absorbed surplus labor force and also the renewal of

aging equipment in the reconstruction improved productivity.

External aid also alleviates economic loss. The region which provides the aid suffers

opportunity loss. Owing to the foregoing positive effects, it is possible that no recognizable consequence follows after an earthquake as far as economic statistics are concerned.

Figure 1 shows the trend of industry output before and after the Kanto earthquake of

1923. There was a clear reduction of the output. Figure 2 shows the index of industrial

products in the Miyagiken-oki earthquake of 197&. The consequence of the earthquake did

not last as long as in the case of Figure 1. Although the magnitude of the latter is not

as big as the Kanto earthquake, industrial and commercial facilities in the Sendai and

400
adjacent areas suffered considerable damage. Outages of electricity, water, and natural

gas supplies also affected industry and commerce to some extent. The reason that the consequences of the earthquake were minor is the prosperity in Sendai due to the opening of

the Tohoku expressway and the construction of the New Tohoku railroad line which generated

new demand and external aid.

Figure 3 shows a schematic tendency of an economic index before and after an

earthquake. Curve (C) is the tendency without earthquake effects. Curve (.A) shows the

tendency assuming that the economic structure after the earthquake would not change compared to that before the earthquake. The drop in curve (A) can be named the loss potential , which is determined by earthquake dimensions, structural resistance, and economic

structure. On the other hand, the actual economic index may follow curve (B) when the

consequences of the earthquake are mitigated. The difference between curves (A) and (B)

can be called the economic resistance , since it is the difference between the anticipated

and actual economic loss. The economic resistance is determined by the following factors:

(.1) Conditions of national and regional economy,

(2) Surplus productivity before the earthquake (labor force, equipment, and

inventory)

,

(3) Recovery and reconstruction, and

(4) External aid.

The damaged economy returns to the original trend after a certain period, this is

called the restoration period . The restoration periods of curves (A) and (.B) are not

always the same.

Economic loss is evaluated by using the loss potential, economic resistance, and the

restoration period. The loss with no resistance is expressed by the area enclosed by

curves (A) and (C). It is determined by the loss potential and the restoration period.

It is considered to be the possible maximum loss, since it expects no surplus productivity

or external aid. It is determined by the earthquake dimensions, structural resistance, and

economic conditions within the damaged region. The economic resistance has the effect of

alleviating the maximum. It is determined by economic conditions in and outside the damaged region. The economic loss is expressed, in other words, by the possible maximum loss

and the economic resistance.

401
DECISION MAKING IN OPTIMIZING EARTHQUAKE DISASTER MITIGATION INVESTMENT

Decision making in earthquake hazard mitigation involves selecting the most effective

means to alleviate the losses expected to the society. To do this the decision maker compares the resources invested and the effect of the alternative policies proposed. The

resources invested involve the cost of strengthening structures, land use management,

research, and inconvenience to the society. The effect includes the reduced physical and

economic loss, and social damage.

It is much easier to compare cost and effect quantitatively when the alternatives have

a certain similarity. Comparing the strengthening of high-rise buildings to single family

dwellings is possible by figuring the cost of material, labor, and the reduced damage in

both alternatives. However, comparing land use restrictions from the hazardous area to

increasing hydrants for fighting fires quantitatively is almost impossible, or maybe senseless. In earthquake disaster we have many cases of the latter type.

The most popular method of comparing invested resources to effects is the cost-benefit

analysis. The cost is relatively easy to define, except for the inconvenience to the

society. It is defined as the expected expense of investment. As for the inconvenience to

the society, one example is the unwilling relocation of residents enforced by a land use

program. Some parts can be quantified while most, such as an increase in human comfort,

are hard to quantify. Benefits are rather hard to define. Most narrowly, they are defined

as reduced damage. However, more widely, they involve reduced casualties, increased comfort, and reduced need for disaster investment, which may be difficult to evaluate.

In cost-benefit analysis, investments whose C/B (ratio of cost to benefit) is greater

than I. (J are usually not adopted. In earthquake disaster mitigation, many alternatives may

have a greater ratio than 1.0. However Oppenheim [b] suggested those are not always

rejected. In his "risk aversion" theory the utility to be saved by an earthquake engineering investment is sometimes not proportional to the value in money to be saved. For

instance the utility of the saved money of $10,UUU is not necessarily ten times that of

S1,00U. Following this theory the investment with greater C/B may be aaopted, because it

may have greater utility compared to the cost.

402
It is possible to use this procedure as a tool for decision making. The problem is

the difficulty in determining the utility function. Utility is a subjective matter which

is difficult to translate in a quantitative manner.

Final decisions are not always based on quantitative data. Rather, many decisions

are made using the qualitative information. It is vain or impossible to quantitize all

factors and to compare their values.

How can the logical decision be reached considering all quantitative and qualitative

information? In earthquake disaster mitigation investment, a decision maker has to consider many factors in multi-dimensional planes. It is not appropriate to translate them

into one dimension. Taking property loss, economic loss, and social damage as an example,

they have independent influences on a decision maker (see Figure 4). For example, social

damage is not willingly accepted in order to reduce economic loss. Each dimension of the

information has to be examined independently. The decision maker synthesizes independent

factors and reaches a final decision. The procedures of selecting appropriate dimensions

and of examining them accurately should be further studied.

CONCLUSIONS

(.1) In earthquake disaster mitigation planning, it is recommended that earthquake

damage be examined using the following three items, (.a) physical loss, (b) economic loss,

and (.cj social damage.

(.2) The loss ratio is one of the convenient indexes to evaluate the property loss.

(.3) The economic loss is evaluated by the possible maximum loss and the economic

resistance-.

(4) In making earthquake disaster mitigation investment decisions, physical,

economic, and social factors should be examined multi-laterally

.

403
REFERENCES

Kuribayashi, E. and Tazaki, T. , "An Evaluation Study on the Distribution Characteristics of Property Losses Caused by Historical Earthquakes," lUth Joint Meeting of

UJNR, May, 1978.

Haas, J. E. , et al., "Reconstruction Following Disaster," Cambridge, MA, MIT Press,

1977.

Friesema, H. P., et al., "Aftermath," Sage Publications, 1979.

Kato, T. , "Earthquake and Economy," Earthquake , University of Tokyo Press, 1976 (.in Japanese)

.

National Academy of Sciences, "The Creat Alaska Earthquake of 19b4," human Ecology

Edition, Washington, D.C., 197U.

Oppenheim, I. J., "Economic Analysis of Earthquake Engineering Investment," 2nd

National Conference on Earthquake Engineering, Stanford, 1979.

404
Table 1 RESTORATION OF ELECTRIC POWER

Earthquake Date M Damaged Area 50% 80% 100%

Restored Restored Restored

Kanto

Niigata

To kachi-oki

1923. 9.

1

1964.6. 16

1968.5.16

7.9 Tokyo, Yokohama 10 days 1 month 2 months

7.5 The area with V or more 3 hours 12 hours 1 day

in JMA* intensity scale

(excluding Niigata City)

Niigata City 1 day 3 days

7.9 The area with V in JMA 5 hours 1 day

intensity scale

24 days

2 days

Miyagiken-oki 1978.6.12 7.4 Miyagi Pref. 8 hours 16 hours 1 day and 13 hours

* JMA : Japan Meteorological Agency

Table 2 RESTORATION OF GAS SUPPLY

Earthquake Date M Gas Co. Number of 50% 80%

Customers Restored Restored

100%

Restored

Kanto 1923.9.1 7.9 Tokyo Gas Co. 251,500

Niigata 1964.6.16 7.5 Hokuriku Gas Co. 45,000

Aomori Gas Co. 4,000

Tokachi-oki 1968.5.16 7.9 Hachinohe Gas 4,600 Co.

Towada Gas Co. 1,400

Miyagiken-oki 1978.6.12 7.4 Sendai Municipal 135,863

Gas Bureau

2 months (Unburned

Area)

6 months (Burned

Area)

3 months 5 months 6 months

— — 6 days

2 days 5 days 20 days

12 days 20 days 1 month

12 days 15 days 27 days

405
Fig. 1 Impact of Kanto Earthquake of 1923 on Industries

Source: Statistics of Industries, Ministry of Trade and

Industry, Japan

Number of Factories

Osaka

140

cr> cri cyi cr> cji <y>

Turnout

_j i i i i i i_

c\j co t m to r— oo cm cm cm cm cm cj

CT> CT) CD CT> Oi CJl Oi

* Tokyo (1922) = 100

406
1975 = 100

j i i i i i 1 1 i i i i

1 2 3 4 5 6 7 8 9 10 1 1 12/ 1 978

tEarthquake

Fig. 2 Index of Industrial Product in Miyagi Prefecture

before and after the Earthquake

407
Restoration Period

| Time

Earthquake

Fig. 3 Economic Resistance in Economic Loss due to Earthquake

Social Damage

Economic Loss

Physical Loss

Fig. 4 An Example of Factors to be Considered in Earthquake

Disaster Mitigation Investment

408
WOOD SHEATHED DIAPHRAGMS AS LATERAL FORCE RESISTING ELEMENTS

by

Edwin G. Zacher

INTRODUCTION

Wood sheathed diaphragms have been used in the design of lateral force resisting

systems for less than fifty years in the United States. The performance of some masonry

walled structures with diagonally sheathed wood floors and/or roofs during the 1933 Long

Beach, California earthquake, encouraged some engineers to explore the possibility of using

these building subsystems for resisting seismic forces. Tests were conducted using actual

floor or roof systems of buildings being demolished. The tests led to proposed methods of

analysis for diagonally sheathed diaphragms and acceptance of allowable values for the capacities of them. The draft of the Applied Technology Council project ATC-7, "Guidelines for

the Design of Wood Sheathed Diaphragms" will have an extensive bibliography listing papers

on these and other diaphragm tests. The "Guidelines" is due to be published in four to six

months and it would be inappropriate to include this information in this short paper.

Since these initial tests there have been a number of tests conducted on diaphragms

having dimensions of 16 to 2U feet by 4b to bU feet. The early tests were made on diaphragms sheathed with one inch and two inch thick boards laid at an angle of 4i> degrees or

9U degrees to the supporting framing. There was also one test made on a system with 3 inch

decking. There were different fastening devices ana different details for end posts and

"chords" used in the tests to establish criteria for design of these systems.

The advent of plywood for the construction of floor and roof systems, and the labor

savings made possible with this material, supplanted lumber (board) sheathing in wood frame

construction. The plywood manufacturers association recognized the need to establish the

acceptability of plywood sheathed floor and roof systems as diaphragms. They conducted a

series of tests of plywood sheathed diaphragms of the same size as previously described.

The results of these tests have been used to establish allowable diaphragm capacities and

design procedures for these systems.

409
DESIGN PROCEDURES - THE GIRDER ANALOGY

Plywood sheathed diaphragms can be conveniently analyzed by the girder analogy shown

graphically on Plate 2.1 taken from the "Guidelines".

The "flange", often termed "chord" of the diaphragm can be the double top plate of a

wood frame wall. The "flange" can also be a continuous wood ledger attached to a concrete

or masonry wall. Steel ledgers, usually angles, have also been used. The reinforcement

in the concrete or masonry wall is often used as the "flange".

The flanges of the diaphragms are typically considered in design to provide all of the

resistance to the flexural stresses in the diaphragm. Based on this assumption the force

on the flanges of the diaphragm can be determined by dividing the moment due to the loading

by the distance between the center-lines of the flanges.

The flanges of most diaphragms, due to the length of these systems and the limitation

on the length of available flange material, will have to be spliced. Splices should be

located as far from the position of maximum moment as the length of available flange

material will permit. Metal splice plates are often used with wood flange members. An

oversight in the design of these splice plates which has often been noted is the failure

to check the compression capacity of the plates.

The plywood sheathing is the analogous web of the diaphragm and is assumed to resist

the shears due to the forces acting on the diaphragm. The diaphragms being considered in

this paper are classified as "blocked". Unblocked diaphragms have low capacities and are

not analogous to girders. A major feature of the diaphragm web is the number of splices

required due to the limitation on the dimensions of the plywood panels in relation to the

total dimensions of the system. There are a number of panel arrangements that can be used.

Two of these arrangements are shown on Plate 2.2. The splice in the direction of the

applied forces can be continuous or interrupted. The splices are made to occur over the

framing members or "blocking". The blocking is in effect a splice plate and may take several forms. The arrangement at the top of Plate 2.2 shows interrupted splices and the connections (nails) required to effect the shear transfer are less than are required for the

arrangement at the bottom of the plate. The codes recognize this by requiring that the

nailing along the panel edges for the continuous splices conform to the nailing required

for the "boundary" edges at the diaphragm reactions.

410
Most diaphragms have openings for various purposes, such as duct openings, skylights

and stairways. Small openings which do not interrupt a full panel of plywood are not a

concern in design. Larger openings which eliminate a number of plywood panels are considered in design. The analogous girder is again used for the analysis. Steel girders have

been extensively studied for the effect of openings in their webs. There have been rules

of thumb established for aspect ratios of openings with respect to depth of beam which

indicate whether or not reinforcement of the opening is required. The tests on diaphragms

with openings has been limited and similar rules have not been established for diaphragms.

The Vierendeel truss approach has been used to analyze steel girders with web openings

and is appropriate for diaphragms with web openings. When the opening in the web is centered on the width of the diaphragm the distribution of the shear at the opening is one

half to each portion of the diaphragm remaining. When the opening is not centered the

distribution is made on the basis of the relative widths of the diaphragm remaining on

each side of the opening. Plate 2.6 illustrates the two conditions.

The segments of the diaphragm at each side (ahead of and beyond the opening in the

direction of the applied forces) are analyzed as separate diaphragms for the local forces

acting on them. The forces in the main flanges due to the action of the diaphragm segment

are then combined with the forces due to the action of the total diaphragm. The forces on

the flanges at the edges of the opening are those due to the action of the diaphragm segment only. These forces must be distributed into the body of the diaphragm and requires

that the framing member acting as the interior flange be extended beyond the opening. The

detail for accomplishing the transfer of the flange force will vary depending on the framing arrangement. Metal splice plates are often used with blocking to extend the

discontinuous header joists at these locations.

PERFORMANCE OF DIAPHRAGMS

Buildings utilizing wood sheathed diaphragms have had a variable record of performance

in past earthquakes and wind storms. The performance can generally be associated with the

degree of detail that went into the construction. Where diaphragms have been well detailed

and the details have been followed in construction the buildings have had a good record of

performance. The bad performance record is attributable to poor details for connections or

lack of compliance with the design drawings.

411
The 1971 San Fernando earthquake provided a good test of the detailing and construction practices for connecting wood sheathed diaphragms. There were a good number of one

story warehouse, manufacturing and commercial buildings in the area built during the preceeding twenty years. The change in construction practice during that period was evident

in the differences in damage to the buildings. There were several older buildings with

diagonally sheathed roofs and standard roof joist framing. These had anchorages connecting

the walls to the joists at four foot centers. No damage was observed in connection with

these buildings. A majority of the buildings were constructed with "panelized" roof systems and the wall anchorages were made through ledgers bolted to the walls. Many of these

systems had failures when the connection of the main roof framing was inadequate to

transfer the forces from the walls into the diaphragm.

High winds have resulted in similar performance records for buildings using wood

sheathed diaphragms. The wind forces create a local effect at eaves, ridges, corners and

similar surface discontinuities which causes uplift on the diaphragm sheathing. This local

effect can reduce the effectiveness of the diaphragm although it is not involved in the

diaphragm action. There are fastening devices now available which are capable of resisting

these uplift forces, and thus permit the diaphragm to act as was intended.

DESIGN OF CONNECTIONS

Wood sheathed diaphragms can be designed and constructed to perform more reliably if

care is taken with the connection details. In order to design connection details it is

necessary to understand how these forces are applied. Plates 1.1 and 1.2 illustrate the

forces acting on the various elements of the building. Wind forces acting on the diaphragm

will be those due to the wind on the wall elements tributary to the roof and the horizontal

component of the wind on the roof. The wind forces for which connection details have been

found to be critical are those acting on the walls. Seismic forces are contributed by the

portion of the wall elements tributary to the roof and by the roof itself. In this case,

also, the forces for which the connections have been found to be critical are those from

the walls.

Plates 3.1 and 3.3 illustrate some of the conditions that are encountered in buildings

with diaphragms. The force from the wall in Detail 3.1 A is delivered to the diaphragm

through the sheet metal framing anchor to the blocking, and thence to the diaphragm by the

412
nails through the sheathing to the blocking. The number of nails required to transmit the

force from the blocking to the sheathing can be calculated. The number of nails may be

more than can be safely and prudently installed in one piece of blocking, in which case

additional blocks are installed in adjacent joist spaces and continuity is provided by

sheet metal splice straps. This will extend the distance available to transfer the force

into the body of the diaphragm. A block in a single joist space is not advisable except

for very small tributary forces. Detail 3.1 B illustrates the method of transfer when the

wall is structurally discontinuous, such as at gable ends of roofs. The force from the

wall is transmitted through the sheet metal framing clip to the 2 x - brace, which is

nailed to the blocking, and the force is then transferred to the sheathing as in

Detail 3.1 A. The framing devices used in both details are capable of resisting uplift

forces and would thus provide for vertical and lateral force components due to wind.

Detail 3.1 C illustrates the simplest and most direct method of transferring the forces

from the wall to the diaphragm.

Details 3.3 A and 3.3 B illustrate two different framing arrangements used in this

country for multiple story wood frame construction. "Baloon" framing is illustrated in

Detail 3.3 A. This type of framing arrangement was more prevalent some years ago than it

is today. The "platform" framing arrangement shown in Detail 3.3 B is more common today.

The force from the wall in Detail 3.3 A is that due to the contribution from the wall above

and below the diaphragm. The force is transferred from the wall to the blocking by the

sheet metal twist strap, and from the blocking to the diaphragm by nails through the

sheathing to the blocking. The forces from the wall elements above and below the diaphragms are delivered through separate paths in Detail 3.3 B. The force from the wall

above the diaphragm is transferred by the nails through the sole plate to the blocking.

These nails also penetrate the sheathing, thus affecting the transfer. The force from the

wall below the diaphragm is transferred through the sheet metal shear angle to the block

and is similar to Detail 3.1 A for the remainder of the transfer. Detail 3.3 C is similar

to Detail 3.3 B except that the wall below the diaphragm is concrete or masonry. For this

detail the anchor bolt transfers the force from the wall to the sill. A sheet metal

framing clip is used for transmitting the force from the sill to the block in lieu of the

shear angle used in Detail 3.3 B. This was done to prevent creating a transverse tension

condition in the sill. This condition is prohibited for seismic force resistance.

413
There are often other loads which must be delivered into the diaphragm, such as

concentrated loads from a large sign structure acted on by winds or a heavy piece of

building equipment acted on by seismic motions. Plate 2.10 illustrates the sign condition in the upper portion. The lower portion of the plate illustrates another condition

where large concentrated loads may be delivered to the diaphragm. The forces due to wind

or seismic in the second story portion must be delivered through the second story wall at

the break to the diaphragm, which must transmit this force to the end walls since there is

no interior wall in the first story at the break. The small segment of exterior wall at

this location is inadequate to act as a support. Plate 5.20 illustrates two special loads

being transmitted to the diaphragm. One load is a line load from the penthouse walls and

other loads are concentrated point loads from the tank supports. Adequate vertical support

for these loads must, of course, be provided and, in addition, the lateral forces resulting

from seismic motion must be distributed into the diaphragm. The details at the bottom of

the plate indicate the methods used for this purpose.

The connections discussed to this point were provided to deliver the forces to the

diaphragm. As in all structural systems there is a requirement that reactions by provided

to resist the forces that have been introduced. The reaction components of the lateral

force resisting system usually employed with buildings having diaphragms are shear walls.

The connection details for transferring the reaction from the diaphragm to the shear wall

are as critical to the performance of the building as the connections for transferring the

forces to the diaphragm. Plates 3.5 and 3.6 show various details for accomplishing this

transfer.

Details 3.5 A and 3.5 B show the method of transfer involved with two different joist

support arrangements in balloon framed wood wall construction. The force is transferred

from the sheathing in Detail 3.5 A to the header-joist ledger and from the ledger to the

blocking and thence to the wall sheathing. The transfer in Detail 3.5 B is from the diaphragm sheathing to the horizontal blocking, from the horizontal blocking to the vertical

blocking, and from the vertical blocking to the wall sheathing.

Detail 3.5 C indicates a condition where two diaphragms abut to a shear wall at

different levels. The transfer at the upper level is accomplished by the nailing to the

header-joist through the diaphragm sheathing, then to the double top plate by means of the

sheet metal shear angle and finally to the wall sheathing by the nailing of the sheathing

414
to the plate. The force from the lower diaphragm sheathing is transmitted to the headerjoist ledger, from the ledger to the top plate by nailing through the ledger to the plate

and finally to the wall sheathing.

The transfer details on Plate 3.6 are for use with concrete or masonry walls. Detail

3.6 A has a framing arrangement with the framing members parallel to the wall. The ledger

in this detail is the common member for attachment to the sheathing and the wall. The

nailing of the sheathing transmits the force to the ledger and the bolts transmit the

force on to the wall. The ledger bolts are also subjected to vertical loads and the combination of the loads must be resisted.

Detail 3.6 B is a more involved detail having a steel angle ledger. This might be

employed when there are large vertical loads being supported by the ledger. Wood in cross

grain bearing has limited capacity well below the capacity of the bolt in the concrete.

The force from the sheathing is transmitted to the blocking, then from the blocking to the

ledger angle by bearing against the welded plates. The bolts from the ledger angle to the

wall complete the transfer. The number of pieces involved and the need for close dimensional coordination between the framing layout, the ledger angle fabrication and the bolt

spacing in the concrete relegate this detail to use only when other considerations override

the problems enumerated. The alternate solution shown dashed on the same detail eliminates

a number of the problems but creates another problem; the transfer of forces to the diaphragm for the normal direction requires additional devices.

Detail 3.6 C shows a typical "panelized" roof construction with the wall stopping at

the roof line. The transfer here is from the sheathing to the sill, and from the sill to

the wall by means of nailing and bolting, respectively. There are an unlimited number of

different conditions that can occur and there are additional details in the "Guidelines"

including design examples with numerical results.

There is one additional type of force transfer from the diaphragm to the vertical

resisting element which needs to be discussed. The details shown thus far had the vertical

resisting element at the location of the boundary of the diaphragm but there are many occasions where this condition is not provided. Plate 5.7 illustrates one condition where the

vertical resisting element extends over only a portion of the width of the diaphragm. In

this case it becomes necessary to introduce a member to "drag" the forces from the diaphragm

to the vertical element, a "drag strut."

415
Plate 2.9 show two other configurations of buildings where drag struts must be

introduced to effect the transfer of lateral forces from the diaphragm to the vertical

resisting elements. The top portion of the Plate shows a type of residence that experienced considerable damage in the San Fernando earthquake. The failure to provide drag

struts and vertical transfer members resulted in essentially total loss of some of these

buildings. The building at the bottom of the Plate illustrates a more complex version of

the conditions shown at the top of the Plate. The problems created by site and architectural arrangement conditions often tax the ingenuity of the design engineer to provide a

structural system that will provide the required resistance to lateral forces.

There are a number of additional configurations and conditions which are encountered

in buildings with diaphragms but this short paper cannot include illustrations and discussions on all of these. Diaphragm stiffness has an important effect on the performance of

buildings. Procedures have been developed for calculating the stiffness of plywood

sheathed diaphragms which have been shown to be somewhat reasonable by test measurements.

Diagonally sheathed diaphragm deflections are not as well known. The period of the diaphragm may govern the response of the building to seismic motions and, since the period of

the diaphragm is related to its stiffness, it is important to be able to reasonably determine the stiffness of this element. Recent strong motion instrumental records will assist

in determining the effects of diaphragm action. Diaphragms are often irregular in shape,

as was shown on Plate 2.10, and special detailing is necessary to resist the stresses

created by the discontinuities. Diaphragms may also be continuous over a number of spans.

The distribution of the forces to the various reactions should be determined on the basis

of the relative rigidities of the diaphragm and the vertical elements of the lateral force

resisting system providing the support.

Wood sheathed diaphragms can be used as parts of lateral force resisting systems.

Buildings utilizing wood sheathed diaphragms will perform well when they are properly

detailed and constructed. The importance of providing continuous paths for transmitting

lateral forces from their point of origin to the final point of resistance at the foundation level can not be overemphasized.

416
I FOOT SEGMENT OF BUILDING

Wind forces in pounds

per lineal foot, plf

ONE STORY BUILDING

WIND FORCES Plate

417
Seismic "pressures" in

I FOOT SEGMENT OF BUILDING

Seimic forces in pounds per lineal foot, plf

ONE STORY BUILDING

SEISMIC FORCES Plate 1.2

418
BUILT UP STEEL GIRDER

Continuous rim joist often used instead - preferable

Blockin

Forces

Joi«t or rafter Range splice

Flange -web

shear transfer

PLYWOOD SHEATHED DIAPHRAGM

GIRDER ANALOGY Plate 2.

419
Panel

Panel edges not continuous in direction

of force

-Blocking

Diaphrogm boundary

CASE

CASE 2

PLYWOOD PANEL ARRANGEMENTS PLATE 2.2

420
SEGMENT OF BUILDING

DIAPHRAGMS WITH OPENINGS PLATE 2.6

421
ONE AND TWO STORY BUILDfNG

SPECIAL LOADING CONDITIONS PLATE 2.10

422
Diaphragm sheathing

Typical nailing

Joist

2 x - blocking

Boundary nailing Nails to blocking

Diaphragm sheathing

16 nails

Sheet metal framing anchor at each block

Stud wall

DETAIL 3.1 A

Gable end stud wall

2x - brace

Anchor bolt

Sheet metal framing clip

DETAIL 3.1 B

Diaphragm sheathing

Roof joist

Boundary nailing

Joist hanger

Blocking Stud wall

DETAIL 3.1 C

CONNECTION DETAILS FOR

TRANSFER OF FORCES TO DIAPHRAGM Plate 3.

42 3
DETAIL 3.3 C CONNECTION DETAILS FOR

TRANSFER OF FORCES TO DIAPHRAGM PLATE 3.3

424
Stud wall (balloon framing

Typical nailing

Wall sheathing

Blocking

Diaphragm

sheathing

Boundary nailing

Joist

Header joist - ledger

I6 Unails DETAIL 3.5 A

-Boundary nailing —-Stud wall (balloon framing)

-Diaphragm sheathing

I6 d nails* •Typical nailing

I x - ribbon

Blocking _ DETAIL 3.5 B

Boundary nailing

Diaphragm sheathing

Sheet metal shear transfer angle

Joist hanger

Typical nailing

Joist Stud wall Header joist - ledger—

DETAIL 3.5 C

CONNECTION DETAILS FOR

TRANSFER OF FORCES FROM DIAPHRAGM Plate 3.5

425
Boundary nailing

Diaphragm —

sheathing

Concrete (or masonry) wall

Diaphragm

sheathing

mate - 3x- ledger

bolted to wall and joists cut

2 x - blocking

Angle ledger with anchor

bolts

Boundc

Plate welded to angl

bolted to joist DETAIL 3.6 B

Diaphragm sheathing

Joist hanger

Sub - purlin

Purlin

Joist hanger

Concrete (or masonry.

^Sill with anchor bolts DETAIL 3.6 C CONNECTION DETAILS FOR

TRANSFER OF FORCES FROM DIAPHRAGM PLATE 3.6

426
Wind pressures in pounds

per lineal foot , plf

OpenShear wall

5.7A-

-Reaction-fi X X

x

Open

Roof rake overhang

PARTIAL BUILDING

Diaphragm sheathing

Blocking

Splice plate with 4- Ig in. x 12 in. lag bolts

to beam and

4- Ig in. machine bolts

to plate

Splice plate with 8- Ig in. x IZ in. lag

bolts to beam

Beam

WIND FORCES

DESIGN EXAMPLE 5.3

Column

DE TAIL 5.7 B Plate 5.7

427
SPLIT LEVEL RESIDENCE

MULTI LEVEL OFFICE

SPECIAL CONFIGURATIONS PLATE 2.9

428
A PROPOSED METHOD FOR THE EVALUATION OF ASEISMIC

PERFORMANCE OF EXISTING TIMBER DWELLINGS

by

T. Murota

H. Okada

ABSTRACT

In response to strong social concerns about aseismic performance of existing buildings

in Japan, the Ministry of Construction and the Japanese Building Disaster Prevention Association proposed, in 1979, an evaluation method for aseismic performance of existing timber

dwellings. This was in addition to the earlier proposals of similar evaluation methods for

existing reinforced concrete and steel construction buildings of 1977 and 1978. The evaluation method for timber dwellings differs from the preceding two methods in that aseismic

performance can be evaluated not by structural engineers but by owners themselves.

In this paper an outline of the aseismic performance evaluation method for timber

dwellings and some applications are described.

4Z9
INTRODUCTION

In recent years severe earthquakes such as the Izu-Penninsula Earthquake, the

Izuohshima Earthquake, and the Off-Miyage Earthquake have continued to occur. It is predicted that a big earthquake may occur in the near future at a point offshore from the

Tokai area. Under these circumstances a general concern about the aseismic performance of

buildings has been itensified.

Corresponding to this concern, the Ministry of Construction and the Japanese Building

Disaster Prevention Association proposed aseismic performance evaluation methods for existing reinforced concrete buildings, and for steel buildings in 1977 and 1978, respectively.

However, most Japanese buildings are timber constructed dwellings and therefore the above

methods were not enough to satisfy the general concerns. The Ministry of Construction and

the Japanese Building Disaster Prevention Association then began to develop a similar

method for timber construction dwellings in 1978. A committee, which consists of the members named below, was organized and along with the Japanese Building Disaster Prevention

Association undertook the development and completed the work in 1979. An outline and some

applications of the aseismic performance evaluation method for timber dwellings developed

by the committee are described in the following sections.

COMMITTEE MEMBER LIST

Hideo Sugiyama

Yukihiro Kamiyama

Isao Sakamoto

Tatsuo Murota

Hisashi Okada

Koji Ueda

Haruhisa Kato

Kyoichi Kobayashi

Toshibumi Fukuda

Tamio Oba

Chairman, Professor, Department of Agriculture, Tokyo

University

Professor, Department of Engineering, Waseda University

Assistant Professor, Department of Engineering, Tokyo

University

Building Research Institute, Ministry of Construction

Building Research Institute, Ministry of Construction

Housing Bureau, Ministry of Construction

Housing Bureau, Ministry of Construction

Housing Bureau, Ministry of Construction

Housing Bureau, Ministry of Construction

Urban Planning Bureau, Tokyo Metropolis Administration

430
Kuniaki Shimazaki Urban Planning Bureau, Tokyo Metropolis Administration

Tatsuo Watanabe Urban Planning and Housing Department, Shizuoka Prefecture

Atsushi Maseki Building Safety Association of Aichi Prefecture

Hiroshi Nakajima Japanese Architecture Association

Mikio Maeoka Japanese Building Disaster Prevention Association

ASEISMIC PERFORMANCE EVALUATION METHOD

The purpose of this aseismic performance evaluation method is to provide residents

with safety measures for their houses. Using this method, residents may easily find conclusions to the problem of whether their houses can better resist strong earthquake

motions.

In the method, residents are asked to judge the following six items:

(a) soil condition and type of foundation

(b) weight of roof and number of stories

(c) arrangement of walls

(d) existence of diagonal members

(e) length of walls

(f) extent of aging

Residents can make the judgment easily by drawing a first floor plan of their houses

(Figure 1), making observations at several places in their houses, and by then making

simple calculations.

Drawing the First Floor Plan

As the first step residents are requested to draw the first floor plan on graph paper.

The first floor plan is necessary to calculate floor area, wall length, and to determine

wall arrangements. An example is shown in Figure 1. The length unit for the drawing is

the "ken" which is a traditional Japanese unit of about 1.8 m.

Calculation of Floor Area

The calculation of floor area is to be made in a Japanese unit called the "Tsubo"

(= 1 ken x 1 ken =3.3 m

z

).

431
Completing the Mark (or Ranking) Sheet

Following Table 1, as the next step, residents are to select appropriate marks for

each items from a to f described below which correspond to results of observations or

calculations requested.

In cases where more than two marks may be selected for an item, the smaller values are

to be selected.

Procedures for making the judgments or calculations and the background for the ranking

scheme shown in Table 1 are as follows:

(a) Soil Condition and Type of Foundation

Soil condition is closely related to the acceleration level put into houses

during an earthquake. Footings of timber houses are generally embedded into very

shallow ground, and this often leads to differential settlement of the surface

ground. Damage is observed more in the case of separated footings than in the case

of continous footings. These facts are reflected in the marks for item a.

Classification of soil condition is generally very difficult for nonprofessionals

and therefore the following suggestions are provided for reference:

hard ground — rock, sand gravel, gravelly soil

soft ground — marsh, reclaimed land, artificial fill ground, top of slopes

(b) Weight of Roof and Number of Stories

This item is related to the input lateral force caused by earthquakes. In

general, the shear force which acts on houses is roughly proportional to the dead

weight of houses, which consist, mainly of roof weight and number of stories. Marks

for this item are decided by these facts.

Roof weight is to be judged by referring to examples below:

heavy roof — clay-tile roofing, cement-tile roofing

light roof — iron-sheet roofing, asbesto-cement roofing

The number of stories is confined to two cases. This is because timber construction of more than 3 stories are prohibited as a rule and basements are very rare in

Japan.

(c) Arrangement of Walls

Observing the earthquake damage to timber dwellings, it is noted that the

arrangement of walls and openings on the exterior wall planes has a great effect on

432
the extent of damage. In this evaluation, wall-opening arrangements are classified

into six types as shown in Figure 2.

(d) Existence of diagonal members

Earthquake resistant structural elements in Japanese houses of timber

construction are mainly plaster walls, walls of board siding, and diagonal bracings.

In the Japanese building code, a relative bearing capacity for these elements is

provided, i.e., bearing capacities of board siding walls and diagonal bracings

(30 mm x 90 mm cross-section) are 1.5 times that of plaster walls.

The object of this item is to estimate the bearing capacity of the walls. In

cases where diagonal bracings are adopted, the corresponding mark for this item is

1.5 and in other cases 1.0. The mark for board siding walls can be estimated as 1.5

according to the Japanese code but in this evaluation method 1.0 is adopted because

the value 1.5 seems to be over-estimated considering the test results.

Usually the existence of diagonal members can not be determined from external

appearance, therefore, except when the existence is confirmed by design documents or

memory it must be judged that diagonals do not exist.

(e) Length of Walls

Referring to the first floor plan drawn before, the length of walls are to be

summed up for two perpendicular directions, respectively. The lesser of the two

values divided by the first floor area previously calculated determines the mark for

this item according to Table 1.

In the Japanese building code, a minimum requirement of wall length/floor area i

provided. According to this provision, in cases where a one-story timber house is

constructed on continuous footings and plaster walls are adopted as earthquake resistant structure elements, the total sum of the plaster wall length in a direction must

be more than 15 cm per unit floor area 1 m^. Reducing the unit into "ken/tsubo"

which is used in this evaluation method, it becomes 0.275 ken/tsubo.

From this fact a mark of 1.0 is taken as a standard for this item when the wall

length/floor area is in the range of 0.275 + 0.025. Therefore a mark of 1.0 for this

item is considered to be legal concerning the wall length requirement for earthquakes

433
(f) Extent of Aging

This item is provided to estimate the decrease in bearing capacity which may have

been caused by aging. Marks for this item are based on the assumption that when

things like deformation to structural frames or foundation, cracks in footings,

remarkable decay in timbers, nesting of white ants, etc. are seen, aseismic performance

may have decreased.

Total Marks and Judgment

The total mark is obtained by multiplying all the marks for items a to f. According

to the value of the total mark, residents can arrive at one of four judgments as shown in

Table 2.

For a house with a total mark less than 0.7, the house is judged to have a possibility

of collapse, and it is recommended the owner consult with structural engineers about retrofitting. Houses whose mark is 0.7 to 1.0 are judged not to have enough aseismic performance

and a more detailed inspection is recommended. When the mark is more than 1.0, it is

judged that the house will be safe in earthquakes but when the mark does not exceed 1.5 it

is still thought to be better to ask a structural engineer to make an inspection. This is

due to the calculation of wall length where all walls are considered effective when some

walls may, in fact, be ineffective as structural members.

APPLICATION

Results of the application of this evaluation method to existing timber houses are

shown in Table 3. Sixty-two houses were selected as objects of application. No. 1 to 50

are houses in Kodaira City, Tokyo; No. 51 to 53 are those in Shizuoka Prefecture; No. 54 to

60 are those which appeared in the magazine "New Housing;" and No. 61 and 62 are those of

"daikoku-bashira" construction, which is a traditional Japanese construction.

The distribution of total marks for these 62 evaluations is shown in Figure 3. Mean

and standard deviation of total marks were 1.56 and 0.51 respectively. Only one house

(1.6 percent) scored less than 0.7 and was judged liable to collapse. Five houses (8.1

percent) scored 0.7 to 1.0 and were judged to need a more detailed inspection. Fifty-six

houses (90.3 percent) scored more than 1.0. Thirty-four houses (54.8 percent) ranked

higher than 1.5, and they were judged to have sufficient aseismic performance.

434
CONCLUDING REMARKS

In this paper the aseismic performance evaluation method for timber dwellings proposed

by the Ministry of Construction and the Japanese Building Disaster Prevention Association

is described. Many problems concerning the accuracy of results obtained using this method

can be seen. The purpose of this evaluation, however, is to provide large numbers of residents with a rough measure of the aseismic performance of their dwellings', and it is considered almost inevitable that accuracy will be sacrificed to some extent.

There is much interest in the predictive accuracy of this evaluation method. However,

no data has been collected as yet, but as earthquakes occur it is expected that detailed

comparisons will be made.

435
Table 1 Marks for item a to f (Ranking Criteria)

Mark for item a soil condition and type of

>X ^ >N> foundation

type of foundation soil condition

hard medium soft

continuous

concrete footing

others

Mark for item b weig

of s

1.2 1.0 0.7

0.7 0.6 0.5

ht of roof and number

tory

number of story weight of roof

heavy light

1

2 1.3 1.0

0.7 0.6

Mark for item c arrangement of walls

type of wall arrangement mark

A 1.2

B 1 . 0

C 0.9

D 0.8

E 0.7

F 0.8

Types of wall arrangement are shown in Fig.

Mark for item d existence of diagonal

members

diagonal members mark

exi st 1.5 none 1.0

436
Table 1

Mark for

continued

item e length of walls

total wall length mark first floor area

L/A < 0 .08 0 . 2

0 . 08 < L/A < 0 .lU 0.U

0 . Ik < L/A < 0 .19 0.6

0 . 19 ^ L/A < 0 .25 0.8

0 . 2 5 =£ L/A < 0 .30 1 . 0

0 . 30 <. L/A < 0 .35 1 . 2 0.35^ L/A < 0 .Ul 1 . 4 0 .Ul < L/A < 0 .U6 1.6

0 . U6 ^ L/A < 0 . 52 1.8

0 . 52 ^ L/A < 0 .57 2.0

0 . J i S L/A < 0 .63 2.2

0 . 63 < L/A < 0 .68 2.h

0 . 68 ^ L/A < 0 .71* 2.6

0jll<L/A<0 .79 2.8

0 . 79 < L/A 3.0

Mark for item f extent of aging

extent of ag ing mark

not decayed 1 . 0 decayed 0.9

cracks in foundation 0.9

eaten by white ants 0.8

Total mark P

P= axbxcxdxexf

Table 2 Judgement of results

total mark judgement comment

P < 0.7 may possibly be collapsed should consult with structural

engineers about retrofitting

0.7 <P <1.0 necessary to make more detailed inspection

ask structural engineers to make an inspection on site

1.0 <P <1.5 safe better to ask structural

engineers to make an inspection

1.5 ^P safe

437
Table 3 Result of application

No. number

of

story

first

floor

area

(m 2

)

marks of each item

total mark

a b c d e t

1 2 11.6 1.0 0.7 1.2 1.5 2.0 1.0 3.02

2 2 14.0 1.0 0.6 0.8 1.5 1.6 1.0 1.38

3 2 27.4 1.2 0.6 0.8 1.5 1.0 1.0 0.86

4 2 15.8 1.0 0.6 1.0 1.5 1.8 1.0 1.94

5 2 15.4 1.0 0.6 1.0 1.5 1.8 1.0 1.94

6 2 13.4 1.0 0.6 1 .2 1.5 1 .8 1.0 2.33

7 2 20.8 1.0 0.6 1 .0 1.5 1 .6 1.0 1.73

8 2 14.2 1.0 0.6 1 .0 1.5 1.3 1.0 1.94

9 2 12.7 1.0 0.6 0.9 1.5 1.6 1.0 1.56

10 2 8.8 1.0 0.7 0.9 1.5 1.0 1.0 1.14

11 2 8.7 1.0 0.7 0.9 1 .5 1.4 1.0 1.58

12 2 10.6 1.0 0 7 1 .0 1 .5 1 .2 1.0 1.51

13 2 12.5 1 .0 0 6 0 ft 1 .

5

i a 1.0 1.56

14 2 35.9 1.0 0.6 0.8 1 .5 1 .2 1.0 1.03

15 2 11.1 1.0 0.6 0.9 1 .5 1 0 1.0 0.97

16 2 26.2 1.0 0.6 0,8 1 .5 1 .4 1.0 1.21

17 2 23.8 1.0 0.8 1 .5 1 2 1.0 1.03

18 2 11.5 1.0 n 7 n ft 1i • Tj 1 ft 1.0 1.81

19 2 20.0 1.0 0.8 1 .5 f) ft 1.0 0.70

20 2 8.7 1.0 n 7 l n 1 ft 1.0 2.26

21 2 11.0 1.0 0.7 1.2 1.5 1.8 1.0 2.72

22 2 10.3 1.0 0.7 0.8 1.5 1.6 1.0 1.61

23 2 10.5 1.0 0.7 1.0 1.5 1.6 1.0 2.02

24 2 18.2 1.0 0.6 0.8 1.5 1.2 1.0 1.03

25 2 9.5 1.0 0.6 0.9 1.5 1.4 1,0 1.36

26 2 7.7 1.0 0.7 0.8 1.5 1.4 1.0 1.40

27 2 11.0 1.0 0.6 1.0 1.5 1.6 1.0 1.73

28 2 11.0 1.0 0.6 1.0 1.5 1.6 1.0 1.73

29 2 10.1 1.0 0.6 0.8 1.5 1.4 1.0 1.21

30 2 12.1 1.0 0.6 0.9 1.5 1.6 1.0 1.55

31 2 11.3 1.0 0.6 0.8 1.5 1.6 1.0 1.38

438
Table 3 continued

No. number

of

story

ti rst floor

marks of each item

total area mark

\m ) a b c d e T

32 2 11.6 1.0 u • o 1 c 1 . D T ft 1.0 1 .56 33 2 11.3 1.0 n fi i1 • nu I * J 1 ft 1.0 1 .94

34 2 12.1 1.0 n k 1 • u 1 J T ft 1 • o 1.0 1 .94

35 2 11.5 1.0 n ft 11 . 3 1 . u 1.0 1 .38

36 2 14.4 1.0 nU fi

. 0 i n 1 C

1 * J 1.0 1.51 37 2 16.5 1.0 n 7 1 • 3 1 9 1.0 1.51

38 2 9.5 1.0 n 7 i n 1 <i

1 • J 1 .0 2.02

39 2 11.2 T.O u . 0 I • u 11 * 95 1 • O 1.0 1 .73 40 2 11.3 1.0 n 7 i n 1 . u f • 3 £ . u 1.0 2.52

41 2 9.0 1.0 n 7 u . / n q 1 • 3 1 ft 1.0 2.04

42 2 9.4 1.0 u . 0 n ft U . o 1 • 3 i # H 1.0 1.21 43 2 42.4 1.0 u . 0 n ft

1 . 3 1 • f 1.0 1 .03

44 2 13.5 1.0 U . 0

i n 1 . u 1 . D c . U 1.0 2.16

45 2 11.2 1.0 u . 0

i n 1 . 39 n 1.0 2.16

46 2 13.7 1.0 u . 0

n ft

1 . J

l ft

1 . 0 1.0 1 .56

47 2 6.8 1.0 n 7 n q 1 . D

9 9 1.0 2.50

48 2 10.2 1.0 U . O n q 1 . 3

1 9 1.0 1 .03

49 2 10.0 1.0 u , 0

i n 1 . u 1 . DT 9 1.0 1,30

50 2 13.1 1.0 U . 0 1 . u 1 . 3

1 ft 1.0 1 .94

51 1 23.6 1.0 U . O l . U 1 . D 1 . t 1.0 1 .26

52 1 27.8 1.0 1 .0 0.8 1 .5 1 .6 1 .0 1 .92

53 1 25.9 0.7 0.7 0.8 1.5 1 .4 1.0 0.82

54 2 11.4 1.0 0.6 1.0 1.5 1 .8 1 .0 1 .62

55 2 14.7 1.0 0.6 0.9 1.5 1 .8 1.0 1 .46

56 2 20.0 1.0 0.6 1.0 1.5 1.6 1 .0 1 .44

57 2 23.7 1 .0 0,6 1.0 1.5 1.4 1.0 1 .25

58 2 15.3 1 .0 0.6 0.8 1.5 1 .4 1.0 1 .01

59 2 19.0 1 .0 0.6 0.9 1.5 1.4 1.0 1.13

60 2 10.5 1 .0 0.6 1 .0 1.5 1.4 1.0 1 .25

61 1 14.0 0.6 1 .0 0.8 1 .0 1 .6 1.0 0.77

62 1 22.4 0.6 1.0 0.8 1 .0 0.8 1 .0 0.33

439
lK=l80cra X shows invalid walls

of less than 0.5K

in lencth.

Fig. 1 Drawing example of first floor

plan .

corner

(A)L-shaped pair of vails

are arranged on every corner of first floor

plan

.

post

Cc)More than a corner are not stiffened by walls

(B)L-shaped pair of walls

and single walls are arranged on every corner

of first floor plan.

open

(D)One side is opened

clearly

.

open

(E)Two sides are opened

clearly.

more than

3 . 6m

(F)Openings continue more than 3.6m in length.

Fig. 2 Types of wall arrangement

t4()
COID

COo

Xi

CmO

u

<u X)s

Fig

15 ••

10 •

5

0

mean

0 0.5 10 1.5 2.0 2.5 3.0

total mark

3 Frequency distribution of total marks

441
AN EVALUATION METHOD ON THE ASEISMICITY OF EARTH STRUCTURES (ROAD EMBANKMENTS)

by

Tatsuya Fujii

Yasushi Sasaki

ABSTRACT

This paper describes a simple method of aseismatic evaluation of earth structures

(road embankments).

With the exception of large scale structures, few earth-filled structures have been

constructed so far taking aseismicity into consideration.

However, in view of the anticipated occurrence of large-scale earthquakes, it has

become necessary in some parts of Japan to conduct surveys on the aseismicity of existing

earth-filled roads to reduce the danger to road traffic as well as to maintain transport

functions after the earthquakes have occurred.

Though this survey is necessary for other road structures, such as bridges, tunnels,

etc. as well, this paper confines itself to earth-filled roads and presents a method of

aseismatic evaluation based on analysis of damage to earth structures from past

earthquakes.

442
0

INTRODUCTION

There are many examples of disasters and damage involving Japanese earth-filled

structures, such as road embankments, river dykes, and filled railroad beds which were

caused by large-scale earthquakes which have occurred in the past.

Up to the present, most earth-filled structures have been constructed without any

estimating and testing in advance of their seismic strength or their earthquake-resistant

abilities, except for the important large-scale structures such as earth-fill dams. The

reason for this is that functional recovery after damage to such structures made of filled

earth is considered to be relatively easy, since the earth itself is used as reconstruction

material. In addition, appropriate earthquake-resistant designs have not been established.

To establish an appropriate earthquake-resistant design, it is necessary to clarify

forecasting methods for the magnitude of seismic force taken as the design force, the

deformation and strength characteristics of soil due to external dynamic forces, the pattern of failure, and the amount of deformation to earth structures. Many researchers have

been exploring these questions.

It has also become necessary to evaluate the earthquake-resistant strength not only of

proposed earth structures, but also of existing earth structures. In Japan, additional

measures to prepare for earthquake disasters are increasingly expected by the citizenry-atlarge, as seen in the establishment of new laws requiring more preparation for a largescale earthquake in the Tokai District which is expected to occur in the near future.

To meet society's demands, it has become necessary to develop methods for examining

the earthquake-resistant strength of earth structures, primarily, that of existing earthfilled roads. However, to investigate the aseismicity of existing earth structures, the

soil characteristics of underlying subsoil and the earth fill must be known in some detail.

In reality, such detail has not been collected in most instances.

Thus, it has become necessary to develop a simple method of examination which enables

judgments to be made on the probability of damage. The method described in this paper is

based on information currently available, though it may be extremely limited in quantity

and kind.

It is also based on results selected from cases of damage to earth-filled structures

which were caused by relatively large-scale earthquakes in the past.

443
OUTLINE OF DAMAGE TO EARTH STRUCTURES FROM PAST EARTHQUAKES

In the 1964 Niigata Earthquake, damage to river embankments and earth-filled roads

resulted from the liquefaction of saturated sandy soil. In the 1968 Tokachi-Oki Earthquake, damage to road embankments was intensified because of the reduced strength of soil

caused by rainfall which occurred immediately before the earthquake. Many newly constructed earth-filled railroad beds were also damaged. In the 1975 Ohita-Chubu Earthquake

and the 1978 Izu-Ohshima-Kinkai Earthquake, damage to earth-filled roads in mountainous

regions was severe. In the Miyagi-Ken-Oki Earthquake which occurred in June, 1978, much

damage took place along road embankments and river dykes constructed on soft ground.

Damage also occurred to earth-filled structures for roads which were constructed in

mountainous regions.

Among earth structures damaged by these earthquakes, the structures for which data on

ground conditions and cross sections before and after the earthquakes were available have

been collected and are shown in Tables 1 and 2. Most of these damaged earth structures

are located in regions where the Japan Meteorological Agency seismic intensity is equal to

or greater than V. These damage cases include not only roads but also river dykes and

railroad beds which are similar in type and size to the earth structures for the roads.

In the same manner, selected examples of damage to earth structures constructed on

slopes, such as mountain roads, are shown in Table 2.

Damage patterns to earth structures on flat lands due to earthquakes, as set forth

in Table 1, can be generally classified into the five damage patterns shown in Figure 1.

Type-I damage is a collapse of the slope of embankment. In the case of roads, this is

mostly limited to the slope, without causing damage to traffic lanes on the road surfaces.

Type-II damage is a collapse that has almost reached the center of the top of the

structure. This represents an increase in the scale of failure.

Type-Ill damage is a collapse of an earth structure in which the structure is

destroyed to such an extent that the original shape of the structure cannot be recognized.

This includes the collapse of underlying subsoil. This kind of damage frequently occurs

where the subsoil has liquefied.

Type-IV damage is a deformation where the original shape of a structure is maintained

to a certain degree as the result of uniform settlement.

444
Results of the classification of damage set forth in Table 1 are shown in Figure 2 as

a frequency distribution diagram for each damage pattern.

This figure shows that earth-filled structures without pavement on top, such as most river dykes and railroad beds, fit type-II or type-Ill damage patterns while earth-filled

road structures topped with pavement fit type-I or type-II damage patterns.

Next, the grade of damage to embankments is classified by the following three ranks,

which depend on the degree of hazard to road traffic functions caused by the embankment

damage.

Damage rank a: Serious damage to the whole top portion (the road surface) of the earth

structure. Motor vehicle traffic is almost impossible.

Damage rank 3: Degree of damage is slightly less than a. Ordinary traffic functions

on the roads are not possible, but partial functions, such as singlelane traffic remains.

Damage rank y: No damage, or minor damage. Traffic functions are virtually

maintained

.

Collected damage examples include not only roads but river dykes and railroad beds of

the earth-filled type. Naturally, traffic function impairments do not occur to dykes.

Also, the degree of actual impairment of some traffic functions immediately after earthquakes is not recorded for such earth structures.

Therefore, it was decided to rank the state of deformation to earth structures so they

correspond to the traffic function defects in the three ranks just cited, based on crack

widths and the amount of structural settlement. The standards for such determination are

indicated in Table 3. The standards for judgment are given for each damage pattern to

earth structures on flat land. Damage patterns have not been determined separately for

embankments on a slope.

The degree of damage determined by using this table is shown in Tables 1 and 2.

Figure 3 shows the width of cracks and the amount of settlement damage to earth

structures on flat land for each damage pattern. In this figure, river dykes whose tops

are used as roads are classified in the road category.

This figure shows that the amount of settlement of embankment top for heavily damaged

earth structures such as in type-Ill is greater than 1 meter, and that more than one-half

of the type-I and type-II structures settled more than 0.5 meter.

445
APPROACHES TO THE ASEISMATIC EVALUATION OF EARTH STRUCTURES

The cases of damage to earth structures constructed on flat land shown in Table 1,

cases where ground conditions and fill patterns are known, have been analyzed. It was

found that damage of rank y occurred only to earth structures of less than 6 meters in

height, and structures of more than 6 meters in height suffered damage of rank a or rank y

(Figure 4).

Thus, it may be stated that the degree of damage from earthquakes tends to increase

as the embankment height increases. It is known that factors other than the embankment

height affect the degree of damage, since even earth structures less than 6 meters high

and other structures with gentle slopes suffered damage of rank a or rank 3*

Among factors other than embankment height, underlying subsoil characteristics are

considered the most important.

Figures 5 and 6 show the distribution of damage degree in relation to fill height on

the vertical axis and the mean N value (blow count of the standard penetration test) of

ground on the horizontal axis for underlying subsoils with sandy soil and clay soil,

respectively.

On sandy soil, it was found that severe damage, equivalent to damage rank a,

frequently occurred at places where N values were less than 10.

Even where the embankment height was relatively low, ground where the N value was

less than 6 sometimes suffered rank a or 3 damage.

With clayey soil, the distribution of damage ranks begins to change near the N value

of 4 (Figure 6). That is, if the N value is less than 4, the degree of damage tends to

increase even though embankment height is low. Most rank a damage occurred at locations

with N values of less than 4. Damage on cohesive soil where mean N values exceed 8 is not

included in Table 1.

Earth structure damage cases on slopes shown in Table 2, only those caused by the

Miyagi-Ken-Oki Earthquake were compiled. The results are shown in Figure 7. Only two

cases fit rank a. The distribution of damage in rank 3 and rank y shows that damage tends

to increase as embankment height increases, and as ground slope steepness increases.

Figure 8 contains the cases classified and compiled for each mean slope of the earth

structures shown in Figure 7. It shows the height of earth structures, with mean slopes

446
steeper than 1/1.2, to be at most about 15 meters; this occurs because the bottom portions

of filled earth slopes are confined by retaining walls in such structures. In these cases,

rank 3 damage occurs even when the embankment height is about 5 meters.

With respect to earth structures having a mean slope less than 1/1.5, rank g damage

occurs even when the embankment height is relatively low and the ground slope is gentle.

However, if the data for ground slopes of less than 15 degrees are excluded, the damage

distribution indicates that the degree of damage to structures with mean embankment slopes

gentler than 1/1.5 is almost the same as that of structures with mean embankment slopes

greater than 1/1.2.

For earth structures constructed on ground sloped less than about 15°, it might be

suitable to treat them as the earth structures constructed on flat ground. That is, the

failure surface in such an earth structure is considered to be governed only by failure in

the filled earth, rather than by the boundary surface between the earth fill and the

ground.

The following can be concluded from the above analysis:

(1) Embankment height is an important factor in determining the degree of damage.

(2) In case of earth structures on flat ground, the degree of damage begins to change

at a point where the N value is about 6 for ground of sandy soil, and at a point where the

N value is about 4 for ground of cohesive soil.

(3) For earth structures constructed on sloped ground, the factors which determine

the degree of damage are ground slope and embankment height.

(4) With respect to earth fill accompanied by retaining walls, damage of rank g may

occur, even though the embankment height may be low. Extra precautions will be necessary

when concrete block retaining walls are used, as in the case shown in Figure 8.

(5) If the grade of sloped ground is gentle, the embankment can be treated as if it

were constructed on flat ground.

Based on the facts above, Table 4 shows the method for evaluating the earthquakeresistant strength of existing earth structures which has been proposed.

According to the test methods used in this evaluation, each earth structure will

first be checked to determine whether the embankment is constructed on sloped or flat

ground. Then, ground and fill conditions of each structure will be judged separately to

determine the particular factor, a, b, c, or d, into which each condition fits. An overall

447
rank, A, B, or C, will be established after considering the multiplied effect of such

conditions.

Also, adjustments will be made to the overall ranks, A, B, and C, for embankments with

continuously high water content due to drainage topography, as well as for embankments that

have been disturbed in the past.

The adjusted ranks are defined as follows:

Adjusted rank A: Degree of anticipated damage is high.

B: Degree of anticipated damage is relatively high.

C: Degree of anticipated damage is relatively low.

The standard slope of embankment shown in the Table 4 survey sheet equals these values

as listed in Table 5. Boring data should be used whenever possible in evaluating ground

conditions for embankments constructed on flat lands.

REVIEW OF THE SUITABILITY OF EVALUATION METHODS

This evaluation method has been applied to the actual damage cases listed in Tables 1

and 2. The results indicate that for 54 percent of the cases, ranks evaluated in accordance with the evaluation method correspond exactly to the degree of actual damage caused.

This is shown in Table 6. When those actual damage cases which have degrees of damage one

rank lower than the ranks determined using the evaluation method are included, the rate

becomes 75 percent (Rate II shown in Table 6). If only earth structures with pavement are

considered, rates become 59 percent and 83 percent, respectively. As indicated in Table 7,

the rate of cases for actual damage greater than that evaluated was 19 percent, and the

rate of actual damage less than that evaluated was 27 percent.

Though the above evaluation method is very simple, it is based upon a number of

assumptions; but as indicated by the attached tables, predictions made on the basis of fill

and ground conditions will quite satisfactorily express the degree of damage caused by previous earthquakes. However, many technical points still remain to be solved; for instance,

the after-shocks which follow the main shock frequently occur in actual earthquakes, but

the chance of increased damage from such after-shocks is unclear; and the possibility of

increased damage from post-rainfall earthquakes also is unknown. These points must be

studied further and reviewed again in the future.

448
ACKNOWLEDGMENT

The aseismatic evaluation method described here was developed under the direction of

Dr. Masami Fukuoka of the committee in the Japanese Road Association. The writers are

grateful to the direction of Dr. Fukuoka and to the members of the committee.

449
Table 1 List of Damage Cases of Earth Structures on Flat Ground

Identifi- cation No. Embankment

type Ground

type Soil type Blow countof S.P.T. Embankment

height (cm) Mean Gradeof Embankment

(cotangent

)

Fi 1 1 ing material Damage

pattern

Width of crack

(cm)

Set 1 1 ement

(cm) Damagerank

NG- 1 River dyke m Cohesive 2 3.2 1.9 Sand n 20 70 a

soil NG- 2 IV 2 2 .

7

2.5 n 30 50 NG- 3 2.6 2.2 v Y NG- 4

" IV Sandy silt 5 5.8 2.7 " V — — Y NG- 5 " 3.2 2.1 V — — Y NG- 6 IV Silt 3 3.4 1.9 " m 100 40 a

NG- 7 ii IV 2 3.1 1.8 I 20 20 8 NG- 8 .. IV Clayi sand 2 4.0 1.5 V — — Y NG- 9 ,. IV Sandy soil 8 5.2 1.9 n 7 30 fi NG- 10 „ IV Cohesive 5 6.1 1.9 m — 200 a

soil NG- 11 m Sandy soil 10 3.5 1.9 n 40 60 a

NG- 12 IV Sllty sand 5 3.7 1.8 " n 20 — 6 NG- 13 Railroad IV Sand 5-6 m — 300 a Embankment NG- 14 8.0 m — 200 a

TK- 1 Road IV Clay, sand 3 9.0 1.7 Volcanic n — 100 a Embankment ash TK- 2 " n Sand 3 3.8 1.2 Kuroboku n — 250 a TK- 3 i, IV 3.8 1.7 I — 300 a

TK- 4 5.0 1.7 m — 150 a

TK- 5 " IV Sand, silt 2 2.5 1.3 Sand m — 120 a

TK- 6

., IV Sand, clay 2 - 3 3.0 1.7 Loam IV 5 30 T TK- 7 Ra 1 1 road IV 3.0 m 200 Embankment TK- 8 M IV Clay, sand 2-3 3.5 3.6 Loam m — — a

TK- 9 5.5 n 25 15 8 TK- 10 „ 4.1 Sand mixed

with pumic( n 200 a TK- 11 Railroad IV Clay, sand 2 - 6 3.8 1.7 Volcanic i ' m v — a Embankment

TK 12 0

8 TK- 14 IV Peat Sandy soil n 50 TK- 15 10.0 Volcanic TJ - 600 a

ash, sa nd TK- 16 IV Peat n _ 200 a

TK- 17 „ IV Clay, sand 2-3 9.5 1.5 Sand n 200 a TK- 18 TK- 19 „ IV 12.0 1.8 n 100 a

TK- 20 IV Sandy soil n 100

7.0 n 400 a

5.5 1.4 1 300 R Peat 15.0 1 .

5

500 TK- 24 i 100 6

my- 1 River dyke m Sandy soil 20 3.4 2.1 Sand TJ 5 10 B

MY- 2 m 10 3.2 2.2 n 20 30 B MY- 3 IV 8 4.8 2.7 u 50 25 B MY- 4 m 10 4.2 2.6 u 15 5 6 MY- 5 m 9 6.4 2.9 n 30 10 B MY- 6 m CoheBive 6 5.2 3.8 Clay, silt n 10 25 B soli

IV Sandy soil 5 5.2 3.1 Sand i 60 75 0

i 20 MY- 9 TV

/ ' i 3 MY- 10 IV 12 3.4 3.8 / 1 5 s MY- n IV 3 2 .

8

n 30 30 g MY- 12 IV 10 4.6 1.7 n 30 50 MY- 13 IV Cohesive 5 8.0 2.8 IV 100 B soil MY- 14 IV 4 8.4 2.6 i 50 50 B MY- 15 in 5 R.O 3.1 IV 15 ion B MY- 16 IV 3 5.6 3.4 m

MY- 17 IV 3 4 .0 2.5 in MY- 18 IV 3 5.4 3.5 V Y MY- 19 IV 0 3.2 2.6 m a MY- 20 in Sandy soil 9 2.8 1.8 n B MY- 21 in Cohesive 3 4.8 2.4 Clay, silt u 7 B soil 450
Table 1 (Continued)

MY- 22 " m " 7 5.0 2.9 " H 100 100 a

HY- 23 m 3 4.0 2.5 H 5 5 6

MY- 24 River dyke m Sandy soil 12 4.8 3.2 Sand I 7 0 Y

MY- 25 IV Cohesive

soil

4 5.9 2.3 m

MY- 26 IV 4 3.9 1.6 n 3 1 B HY- 27 in Sandy soil 20 2.8 2.1 m a MY- 28 IV Cohesive

soil

4 8.0 2.8 IV 60 6

MY- 29 IV 3 5.4 3.5 IV B

MY- 30 IV 4 5.4 3 J IV B MY- 31 Road embankment

IV 6-7 1.5 - 1.8 i 50 B

Table 2 List of Damage Cases of Earth Structures on Sloped Ground

Identifi- Embankment Ground

Soil type Ground slope Embankment Embankment Filling Damage Width of Settlement Damage cation No. type type (degrees) height (cm) grade (cotangent) material pattern c rack (cm) rank

NC- 15 Road Embankment

I 34.0 40.0 1.2 Sand 5 15 B

NC- 16 " " Tuf faclus 37.0 15.0 1.5 150 a

20.0 8.4 1.8 300 Loam 7.0 4.0 1.0 - 1.2

16 .

5

4.0 1.0 20 n 8 12.0 1.6 IT 4 breccia

32.0 1.6 20 50

0T- 5 „ Pyroclaatlc

rocks

35.0 35.0 1.2 500 a

MY- 32 " 30.0 12.0 1 .0 2 0 Y HY- 33 " 16.0 7.0 2 0 Y MY- 34 9.0 15 8 HY- 35 13.0 14.5 1.99 20 20 3 HY- 36 16.5 10.6 1.85 0 0 Y HY- 37 25.0 25.6 1.89 300 MY- 38 17.5 26.0 1.85 2 0 B MY- 39 25.0 8.6 1 .44 0 0 Y MY- 40 30.0 34.8 1.53 5 0 B MY- 41 28.5 14.1 1.32 3 0 Y HY- 42 38.5 15.1 0.53 4 0 8 HY- 43 35.0 15.1 0.52 3 0 Y MY- 44 32.5 35.0 1.35 2 0 B MY- 45 30.0 15.2 0.52 0 30 8 MY- 46 31.5 27.1 1.34 2 0 6

MY- 47 25.0 22.7 1.86 1 0 Y HY- 48 29.0 14.1 1.88 1 12 Y

HY- 49 Road Embankment

I 34.0 1 14.7 0.59 0 0 Y

MY- 50 34.5 28.3 2.37 1 0 B

HY- 51 33.5 28.7 1 .34 1 8

MY- 52 18.0 9.0 1.77 2 0 Y

MY- 53 22.5 25.6 1.33 1 0 8

MY- 54 37.5 16.9 2. 10 5 12 B

MY- 55 25.0 21.4 1 .46 5 7 B

MY- 56 23.5 10.5 1 .33 3 0 Y

MY- 57 33.0 25.0 1 .40 3 0 6

MY- 58 35.0 10.6 0.57 1 0 Y

HY- 59 32.0 30.0 1.32 6 4 B

HY- 60 9.5 7.6 1.22 0 0 Y

MY- 61 33.0 10.6 0.53 0 25 B

HY- 62 30.0 19.6 1.32 5 5 Y

MY- 63 21.0 31.5 2.09 2 0 S

MY- 64 32.5 25.4 1.33 25 10 B

HY- 65 12.0 10.9 1.27 0 0 Y

451
Table 2 (Continued)

MY- 66 HY- 67 MY- 68 MY- 69 MY- 70 MY- 71 MY- 72 MY- 73 MY- 74 MY- 75 MY- 76 MY- 77

MY- 79 MY- 80 MY- 81 MY- 82 MY- 83 MY- 84 MY- 85 MY- 86 MY- 87

Road" Embankment

30.0

33.0 43.0

15.5

31 . 5 32.0

30.0

14.5

12.5

10 . 5 31.5 32.0

17. 5

30 . 0 34.0

34.0

24.5 30.0

31.5 25.0

21.0

11.0

26.0

34.2 18.8

14.7

18.1 40.8

23.0

10.7

9.1

9.0

5.0 34.8

16.8 10.8 22.9

24.2 16.8

34.1 20.8 13.0

13.0

10.0

1.51

1.21 1.62 1.33

2.09

1.33 1.36

1.81 1.79 1.78 0.80

1.34

2.10

1.15

1.35 1.80

1.76 1.30

1.24 0.66

1.81

I 96

7

5

120

15

6

30

Table 3. Damage Rank and Degree of Damage

Damage pattern Damage rank

I

6 Width of crack greater than 15 cm, or settlement greater than 20 cm.

Ifat

ground Y Width of crack less than 15 cm, and settlement less than 20 cm.

a Width of crack exceeding 30 cm, and settlement exceeding 50 cm.

nt

on H

6 Width of crack less than 30 cm, or settlement less than 50 cm.

QJB m Q

banl 6 Settlement greater than 50 cm.

EW IV y Settlement less than 50 cm.

V y

No damage

a Embankment completely slipped.

Road surface and shoulder were lost.

n

sloped

grour

6

Cracks and settlement occurred on

the road surface as a result of deformation of the filling. Or cross sectional area for restoration is greater than 2 squre meters.

Embankment

o

y

No damage or hair-line cracks on

road surface not damaged . Or cross sectional area for restoration is smaller than 2 square meters

.

45°
Table 4 Aseismatic Survey Sheet for Embankment

Name of office

Survey taken by Date of survey Year Day:

Embankment

location Name of Line ( ) Distance mark ( km - km). Applicable Bectlon ( km - km)

3 b c d Ground slope l/a " in cotangent a<\ .8 l.SfiaO 3<a<4

Embankment height H - m Embankment Blope 1/x • in cotangent Existence of retaining wall (yes, no)

Embankment with standard slope Hi35m 35>H520 20>H*5 5>H

I?!

las

Is

i 8

Type

Height m Embankment with

slopes steeper than standard slope and approach embankment

to bridge abutment

H>30m 30>HS15 15>Hi5 5>H

Ground

1/4.0 ond

it

ion Blow count of S.P.T. N - Thickness of soft soil layer m Sol 1 improvement (yes , no

)

(Work method: )

Degree of ground

sof tmess

Blow count of S.P.T.

Extremely

snft Soft Others

8SN

Ground

c rBlow count of S.P.T. N - Thickness of loose sand layer m

Depth of groundwater ( m from

ground surface) Classified with old river bed on

land condition classification

map (applicable)

Soil improvement (yes , no

)

(Work method: )

Possibility of liquefaction

Extremely

high Med turn Others /

Embankmer

(Grount Blow count of S.P.T. N<6 65N<10 10=N

li

Emba

Emba nkment heigh t H • m

nkment slope 1/x • in cotangent

Embankment with

standard slope H>10m 10>H£6 6>!I43 3>H Embankment retaining wall (yes, no) Type

Height m Embankment with

sloped ground Hifim 8>HS<i 4>H22 2>H

Embankment on Sloped Ground

Embankment ^^condltlon

Ground

condition \^

a b d

a A II B c

b B B C c

c B C c (

Cross section of Embankment (acale)

Overall rank (A, B, C)

Correction of rank due to notes (1), (2) & (3) shown below. (C -* B , B-*A, A' A)

(1) Water content in filling is always high because embankment is located on watercollection ground.

(2) Settlement or cracks on top of embankment occurred in the past due to rainfall or earthquakes

.

(3) Deformation such as settlement or cracks already appeared in the embankment at Its widened portion after usage began.

Corrected judgement rank (A, B, C)

453
Table 5. Standard Grades for Slopes Corresponding to Filling Material and Embankment Height

Filling material Embankment

height (m) Grade Remarks

Sand with good grading (SW),

Gravel and sand mixed with

gravel (GM) , (GC) , (GW) , (GP)

less than 5 m 1:1.5 - 1:1.8

These are applicable to embankment

on foundation with

sufficient bearing

capacity and without fear of infiltration of water.

A typical soil in Japanese unified

classification cate- gory is shown in

( ).

5 - 15 m 1:1.8 - 1:2.0

Sand with poor grading (SP) less than 10 m 1:1.8 - 1:2.0

Rocks (including muck)

less than 10 m 1:1.5 - 1:1.8

10 - 20 m 1:1.8 - 1:2.0

Sandy soil (SM) , (SC)

,

hard cohesive soil, hard clay

(such as hard cohesive soil of diluvium, clay, Kanto loam)

less than 5 m 1:1.5 - 1:1.8

5 - 10 m 1:1.8 - 1:2.0

Soft cohesive soil (VH

2

) less than 5 m 1:1.8 - 1:2.0

Table 6. Rate of Suitability

All earth structures surveyed (road embankment s railroad beds and river dykes)

Rate of suitability I (7.) Rate of suitability TX (%)

Embankments on flat cohesive ground

55.2 72.4

Embankments on flat sandy ground

kU.k 55.6

Embankments on sloped ground

56.6 83.0

Total 54.0 75.0

Earth structures with pavement (earth-filled roads & river embankments

also treated as roads)

Rate of suitability I (%) Rate of suitability II (%)

Embankments on flat cohesive ground 66.7 77.8

Embankments on flat sandy ground

66.7 88.9

Embankments on

sloped ground

56.6 83.0

Total 59.2 83.1

Rate of suitability:

jj ^ Number of survey points shown by

Total number of survey points

Rate of suitability: Number of survey points shown by Total number of survey points

454
Table 7 Rates of unsuitability

All earth structures surveyed

Rate of having actual

damage greater than

those evaluated in %

Rate of having actual

damage less than those

evaluated in %

Embankments on flat cohesive

ground

13.8 31.0

Embankments on flat sandy

ground

44.4 11.1

Embankments on sloped ground

13.2 30.2

Total 19.0 27.0

Earth structures with pavement

Rate of having actual

damage greater than

those evaluated in %

Rate of having actual

damage less than those

evaluated in %

Embankments on flat cohesive

ground

11.1 22.2

Embankments on flat sandy

ground

11.1 22.2

Embankments on sloped ground

13.2 30.2

Total 12.7 28.2

455
Slippage on sloped surface, failure or settlexent

from cracking is limited to shoulder, thus not affecting the central portion of the top of the embankment (traffic lanes, in the case of road).

H: Slip failure or settlement due to craking in the embankment , occurring at the central portion of the top of the embankment (traffic lanes, in the case

of the road )

.

HE: Collapse also occurs in the underlying subsoil

The original shape of the embankment is unnoticeable.

IV: As a result of uniform settlement of the embankment,

the filling is deformed but embankment shape is maintained to a certain degree.

V: No damage.

Fig. 1 Classification for Damage Patterns

Damage

30 Patterns 10 20

I -111111111111

Paved road

I \ Unpaved road

Earth-filled railroad bed River dyke

Fig. 2 Frequency Distribution of Damage Patterns of Embankment on Flat Ground

456
Fig. 3 Frequency Distribution of Crack Widths and Settlement by Damage Pattern, Embankment on Flat Ground

Width of Crack

I I A Hi Paved nOUnpaved C HB

Ea "h ^V"! D River dyke r railroad beds M

0~< 5 5-<10

10~<15 15~<20

25-<30 30~ Unclear ~f

Settlement

0-< 5 5-< 10 10~< 50 50-<100

100 ~<300

300 ~

Unclear 1 1

0

_1 1 I , I I 10 0

* River embankment used also as road is classified as either A or B.

Fig. 4 Relationship between Damage Ranks, Heights, and Slopes, Embankment on Flat Ground

10 10-

4 -g 4-

03 ©

4

3& -
Relationship between Damage Rank,

Embankment Heights, and Blow

Count N of Ground, Embankment

on Flat Sandy Ground

2-

2 2

, , 1 0 Ol , ,

, r- 10 50 0 2 4 6 8

Frequency Blow count N

| 10

i'

15

458

Fig. 6 Relationship between Damage Rank,

Embankment Heights, and Blow

Count N of Ground, Embankment

on Flat Cohesive Ground
Fig. 8 Relationship between Damage Ranks, Embankment Heights, and Ground Slopes, Embankment on Sloped Land

O CP o

O OO

Ground slope (In degrees) Ground slope (in degrees) Ground slope (In degrees)

(a) Mean slope of Embankment 3 1/1.2 (b) 1/1.2 > Mean slope of Embankment > 1/1.5 (c) 1/1.5 > Mean slope of Embankment

459
WIND-INDUCED VIBRATION OF TRUSS MEMBERS OF THE COMMODORE BARRY BRIDGE

by

Gerald F. Fox

INTRODUCTION

The Tacony-Palmyra Bridge was constructed in 1929 and spans the Delaware River at

Philadelphia, Pennsylvania. The 537-foot main span of the bridge is a steel tied arch with

a trussed rib (Figure 1). Shortly after the bridge was opened to traffic, two of the

longer H hangers broke at the top during a severe two-day storm. The bridge was closed to

traffic and further inspection revealed that other hangers exhibited cracking, some at the

top and others at the bottom. The fractures were due to fatigue resulting from windinduced vibrations. Repairs were made and a horizontal strut added between panel points

4 and 4'. The added bracing has been entirely effective.

The Commodore Barry Bridge was constructed in 1973 and spans the Delaware River

between Chester, Pennsylvania and Bridgeport, New Jersey. It is only 22 miles south of

the Tacony-Palmyra Bridge. It is the longest cantilever truss bridge in the United States

with a main span of 1,644 feet and anchor span lengths of 822 feet (Figure 2). The depth

of trusses is 85 feet at the center of the main span and 230 feet over the piers. The distance between trusses is 72' -6". Many of the web members, diagonals and verticals are H

sections and are unusually light. As is well known, long, light structural H members are

susceptible to wind-induced vibrations which can eventually lead to fatigue fractures

that can be catastrophic.

In March 1973, while the bridge was still being constructed, a storm with recorded

gusts up to 65 miles per hour caused several of the members to vibrate violently. Upon

subsequent inspection, open cracks through the flanges of the H vertical members either at

their top or bottom were discovered. The contractor, Bethlehem Steel Corporation, spliced

the cracked members and installed steel cables to restrain against any additional windinduced problems. These temporary guys proved to be effective and the bridge was completed

and then opened to traffic with the cables still in place.

In the 45 years between the incidents noted for the Tacony-Palmyra Bridge and the

Commodore Barry Bridge there have been many similar failures. Those that have been

recorded in the literature include bridges constructed in Japan, Czechoslovakia, Canada and

460
the United States. In view of the great number of such incidents, one is led to ask the

question, "Why should they continue to occur?" Is there something radically wrong in the

way that engineering information is communicated or are engineers naturally lazy and don't

read the literature available? Perhaps they are not given the time or can't afford the

time to adequately research a project before final design is initiated. Very appropriate

is the old quotation, "He who ignores history is bound to relive it."

In April 1974, The Delaware River Port Authority (DRPA) engaged Howard Needles

Tammen and Bergendoff (HNTB) to investigate the causes and suggest possible solutions for

the wind problems of the Commodore Barry Bridge. Dr. Robert H. Scanlan of Princeton

University served as a special consultant during the investigations.

This paper briefly summarizes the phenomenon known as von Karman vortex shedding that

led to the damaging vibrations of the truss members of the Commodore Barry Bridge. In

addition, the alternatives that were studied to prevent future vibrations from occurring

are described. Also the estimated costs of the various schemes are given.

Von Karman Vortices

Everyone here has, at one time or another, looked down upon a flowing stream from a

country bridge. If you were observant, you noticed the water flowing around the rocks

that protruded above the water and that immediately downstream of and adjacent to these

rocks small vortices or whirlpools formed. You really would be classified as very observant if you noticed that these vortices were not directly opposite each other on either

side of an individual rock, but were rather alternately spaced. This alternate spacing of

vortices given rise to periodic forces perpendicular to the stream flow. Wind acts in the

same manner as flowing water. Wind, as it flows around the truss members of the Commodore

Barry Bridge forms alternating vortices which give rise to forces perpendicular to the

wind-flow direction. This phenomenon was discovered by Professor Theodore von Karman and

is named after him (Figure 3). The frequency of the shedding vortices and therefore the

resulting forces is governed by the following very simple law:

s = fb

V

S = Strouhal number

f = Frequency of shedding

461
b = Width of object normal to the flow

V = Velocity of wind.

Values of the Strouhal numbers used in the investigation are given in Figure 4 and

were obtained from References [1-3].

All structures vibrate from the forces caused by wind vortices. It's only when the

natural frequency of the structural member approaches the shedding frequency that a serious

problem occurs. When these frequencies are equal, a condition of resonance occurs in which

very high amplitudes are reached when low damping is present.

The velocity of wind that would allow resonance to occur can be calculated since the

Strouhal number, flange width and natural frequency of the truss members (Figure 5) in

bending and torsion are known or can be easily calculated. For example, the bending

natural frequency for a beam which is fixed at both ends and subject to a tension force

is:

f = Frequency

C = 3.56 for fixed-fixed condition

E = Modulus of elasticity

I = Moment of inertia

m = Mass per unit length

I = Length of member

P = Applied tension force

P _ = Euler critical buckling load.

Stresses in Truss Members

The resonant wind velocities calculated for the web members of the Commodore Barry

Bridge ranged as low as 32 miles per hour.

The stresses in the members resulting from the resonant wind velocity were calculated

from the following approximate formula from Reference [4]:

f

P cr

462
y = Amplitude of vibration

Wa= Weight of air - 0.0766 lbs per cu ft at 15°C sea level

v = Velocity of wind

b = Object width normal to flow

W]-, = Weight of beam per unit length

f = Natural frequency

S = Logrithmic decrement taken as 0.3 for this investigation.

The stresses associated with bending vibrations were calculated using the above

amplitude displacement. For example, for member U24-L24 the calculated stresses are

shown in Figure 6.

Wind Criteria

Utilizing the methods of H.C.S. Thorn [5], the mean recurrence interval of wind 30 fee

above the ground was plotted (Figure 7). The recurrence interval at the bridge height was

also plotted in accordance with the following formula:

VH = (^-)

Y

x V R HR

where: Vj{ = velocity at a height H

VR

= reference velocity at height HR

y = an exponent depending on the character and/or degree of development of the

terrain. For the bridge site a value of 0.143 was used.

Historical data on the duration of extreme winds are scarce. For this investigation

the plot shown in Figure 8 was assumed [4],

The design wind velocity was established at 100 miles per hour at the height of

bridge which has a recurrence interval of about 50 years. All main members shall be

modified if they undergo detrimental vibrations from winds below the design winds.

Alternative Methods of Performance Permanent Rectifications

There are three types of corrective measures which can be used to increase the

resistance of structural members to wind-induced vibration. These are:

463
- Aerodynamic, in which the periodicity of wind-induced force is destroyed by the

addition of "spoilers" to break up the flow or else the force is eliminated (by

"streamlining" or perforation of the member).

- Mechanical, whereby the energy of the periodic wind forces is absorbed by

mechanical devices and vibration amplitude is thus kept within safe limits.

- Structural, whereby the physical characteristics of the basic member are altered

so as to raise its natural frequency of vibration. This can be done by adding

material to increase elastic stiffness, or by using intermediate supports to

decrease effective length.

For the wind-vibration-susceptible members of the cantilever truss spans of the

Commodore Barry Bridge, the use of the aerodynamic solutions of spoilers and/or streamlining was not considered feasible for reasons of appearance and the considerable lead

time required in order to develop effective solutions by experimentation with various

configurations. Also, there was no assurance that such experimentation would, in the end,

provide a satisfactory solution.

Two basic variants of a structural type solution, the addition of a system of

structural members (sub-struts and braces) and a system using permanent steel restraining

cables, were studied (Figures 9 and 10).

In addition, another structural method which was considered is that of "boxing-in"

those H-section members requiring rectification. This method consists of the welding (or

bolting) on of plates so as to form closed box sections.

For the mechanical solution, a vibration absorber was considered. A tuned vibration

absorber is a system consisting of a small mass attached to the individual bridge members

by means of a rubber stem which provides for both the stiffness and damping required for

the absorber. The natural frequency of the absorber is tuned to the natural frequency of

the truss member. When the bridge member begins to oscillate it excites the absorber which

provides effective damping and thus reduces the amplitude of motion of the bridge member to

safe and tolerable values [6,7].

The costs of the proposed rectifications were estimated for a 100 mile per hour wind

and are as follows:

Sub-struts and braces - $3,230,000

Steel restraining cables - 1,580,000

464
"Boxing-in" of H members - 3,660,000

Vibration absorbers - 1,140,000

In view of their aesthetic appeal and estimated low cost, it was recommended that a

research program be initiated with the following objectives:

(1) To verify by means of laboratory and field testing the effectiveness and

applicability of the mechanical absorber type of solution.

(2) To develop a satisfactory mechanical absorber for use on the structure.

(3) To develop, from the results of previous investigations and the research program,

definite recommendations to DRPA for permanent rectifications.

The research program included analytical investigations, field testing, laboratory

testing and wind tunnel testing. [8]

Analytical Investigations

The analytical investigations included a solution of the response of structural

members equipped with one or more vibration absorbers and a study of the following

parameters in regard to their effect on the vibration response of the member:

Ma/MQ = ratio of mass of absorber to effective mass of member

£a = damping ratio of absorber

^a/ wo

= tuning ratio (the ratio of vibration frequency of absorber to that of the

member

)

The results of these studies are shown on Figures 11, 12, and 13. On these figures

the vertical scale represents the magnification ratio (x/x stat ^ c) . This is the ratio of

the maximum half applitude of the member to the deflection which would result if the force

were applied as a steady load. The damping ratio of the member proper has been assumed as

0.004 throughout.

Referring to Figure 11, which is drawn for an absorber damping of 0.04, it can be

noted that there is little improvement in the effectiveness of properly tuned absorbers

with an increase in the mass ratio beyond about 0.005. For "de-tuned" absorbers there is

little improvement for mass ratios above about 0.015.

Figure 12 shows the variation in absorber performance with a change in damping ratio.

The most important conclusion to be drawn from this figure is that the absorber effectiveness is sensitive to changes in its damping ratio for low mass ratios, but that for a 0.02

465
mass ratio, varying the absorber damping ratio between 0.04 and 0.14 has little effect on

the maximum response.

Figure 13, showing the effect of the tuning ratio on absorber effectiveness, shows

clearly the presence of an "optimum" turning ratio for each mass ratio and again indicates

that absorbers with larger mass ratios are more tolerant of variations in the other

parameters

.

Field Investigation

The field investigation consisted of the three phases of (1) Calibration of synthetic

rubber and wire rope test absorbers; (2) Monitoring the behavior of member 24 with and

without absorbers; and (3) Measuring the natural vibration frequencies of selected members

without absorbers.

Stanford University, California performed the field monitoring and measuring of the

natural vibration frequencies. This was accomplished by means of acceleroraeters mounted

on the members which were connected via a coaxial cable to a Fourier analyzer computer.

The differences between the field observations and theoretical results are shown in

Table 1.

The results of the field investigation led to the following conclusions:

(1) The synthetic rubber exhibited consistent and desirable properties for use in

mechanical absorbers but needed to be larger than 3-1/2" square.

(2) Steel cable is not a satisfactory absorber material due to the high fraction of

critical damping and the inconsistent results.

(3) First mode frequencies of members requiring rectification should be measured in

place

.

(4) The difference in energy levels of the member and the mechanical absorber

indicates probable effectiveness of the absorber. However, wind tunnel tests were needed

to test the results at high wind speeds.

Wind Tunnel Testing

Wind tunnel testing was performed at the Low Speed Aerodynamics Laboratory, Canadian

Research Council, Ottawa, under the direction of Dr. R. L. Wardlaw. The truss member

(prototype member) selected for modeling and laboratory testing was member U24-L24 and the

testing program, developed during discussions between Dr. Wardlaw, Dr. R. H. Scanlan and

466
HNTB, consisted of the testing of a 1:3 scale, two dimensional sectional model and a 1:7.5

scale, three-dimensional elastic model. Both models were tested without and with attached

mechanical absorbers.

The results of the 1:7.5 scale elastic model tests in weak axis motion, strong axis

motion and torsional motion are shown on Figures 14 to 16, respectively. A comparison of

the sectional and elastic model results, in general, were satisfactory.

From the results of the experimental and analytical studies performed during the

research program, it was concluded that a system of mechanical vibration absorbers would

provide the necessary protection against undesirable wind-induced vibrations of the truss

members of the Commodore Barry Bridge. The DRPA accepted the recommendation and gave

approval for the preparation of plans and specifications.

Laboratory Testing

While the plans were being prepared, the designed vibration absorber (Figure 17) was

being further tested at the General Testing Laboratories in Kansas City, Missouri. These

tests provided:

(1) The variation of the natural frequency of the vibration absorber for varying

lengths of rubber stem, combined with a variety of weights as anticipated for actual use

on the Commodore Barry Bridge, under a temperature range of 0°F to 125°F. The plots of

frequency vs. length followed the same trend as obtained by field testing at the bridge

site.

(2) The variation in the absorber damping values for the conditions outlined in

(1) above.

Installation

Bids were taken in May 1976 and the United States Steel Corporation was the low bidder

at a price of $1,162,375. The D. S. Brown Company furnished the absorbers to United States

Steel for erection. A total of 920 absorbers were installed on 166 members. The temporary

cables were removed and the installation of the absorbers was completed in August 1977.

Since that time there have been no problems reported and two inspections of the

absorbers have taken place. While the solution of vibration absorbers was correct for

remedying an existing condition of the Commodore Barry Bridge, it does not necessarily

467
follow that absorbers should be used for initial design. It would appear superior to

provide adequate stiffness or sufficient damping, or both, in the original design.

REFERENCES

[1] Transactions of the American Society of Civil Engineers, Vol. 126, 1961, Part II,

p. 1124, "Wind Forces on Structures."

[2] Yamaguchi , T. , Shiraki, K. , Umeraura, S. and Tanaka, H. , "Vibrations Caused by

Von Karman's Vortex on Bridge Members and its Counter Measure."

[3] Otsuki, Y. , Washizu, K. , Ohya, A., Tomizawa, H. , and Fujii, K. , "Experiments on

the Aeroelastic Instability of Prismatic Bars with Rectangular Sections," presented

at Tokyo Wind Conference, 1971.

[4] Merrison Committee Report Great Britain, "Inquiry into the Basis of Design and

Method of Erection of Steel Box Girder Bridges," 1973, p. 4/23.

[5] Thorn, H. C. S. , "New Distributions of Extreme Winds in the United States," Journal

of the Structural Division, ASCE, Vol. 94, No. ST7, July 1968, p. 1787.

[6] Timoshenko, S., "Vibration Problems in Engineering," Third Edition, D. Van Nostrand

Company, Inc., 1955.

[7] Wardlaw, R. L. , and Cooper, K. R. , "Mechanisms and Alleviation of Wind-Induced

Structural Vibrations," Proceedings - Second Symposium on Applications of Solid

Mechanics, McMaster University, Hamilton, Ontario, Canada.

[8] Irwin, H. P. A. H. , Cooper, K. R. , Wardlaw, R. L. , "Application of Vibration

Absorbers to Control Wind-Induced Vibration of I-Beam Truss Members on the Commodore

Barry Bridge," LTR-LA-194, National Research Council of Canada, Ottawa, January 1976.

468
TABLE I MEASURED AND COMPUTED FUNDAMENTAL

FREQUENCIES (HERTZ)

ff.ember Field Observations Theoretical Results*

Weak Strong Torsional Weak Strong Torsional

Members w/o Sway Frames or Intermediate Struts

2h

D 3.^9 h.G7 3-05 3-51 5-31 3.52

2« 4.18 7.10 3.10 Z».15 7.77 3.86

Members with Sway Frames

15 2. 3k -- 3.28 2.24

29 2.58 -- -- 2.39

Members with Horizontal Strut and w/o Sway Frame

20 -- 2.58 -- - 2.8O

20** 3.28 2.58 2.85 3.37 2.80 3-73

* Effect of bracket weight included

!t * 2nd mode frequencies

Tacony- Palm yra Bridge

FIGURE 1

469
Fig. 2 Commodore Barry Bridge

S = ft

v— 1/ e I o £ 1 1 of u/iWd

Figure j

470
H- SECTION

(0.7 ^ %^ 1.2)

BOX SECTION

(0.50^ b

/d^2.0)

BEVELED

CORNER

SHARP

CORNER

©(DEC) S

b

/d S SHARP S BEVELED

0 0.120 0.50 0.079 0.087

20 0.153 1 .00 0.125 0.138

40 0.1 16 1 .50 0.136 0.150

60 0.133 2.00 0.140 0. 154

90 0.137

Transactions of the American

Society of Civil Engineers,

Vol. 126, 1961, Part II, P

u ll24 "Wind Forces on Structures"

Vibration caused by Von Karman's

Vortex on Bridge Members and

its Countermeasure by

T. Yamaguchi, K. Shiraki,

S. Umemura and H. Tanaka.

Source^ Experiments on the aeroelastic

instability of prismatic bars

with rectangular sections by

Y. Otsuki, K. Washizu, A Ohya,

H. Tomizawa and K. Fujii. Presented at Tokyo Wind

Conference, 1971.

STROUHAL NUMBERS (S)

471
MEMBER VIBRATION

FIGURE 5

472
L24

ELEVATION

y33,!7Q psi + 33,170 psi

+ 13,800 psi

13,800 psi

- 33,170 psi

+33,170 psi

BENDING STRESS

WEAK AXIS

HALF AMPLITUDE

+ \ 1,730 psi

AXIAL STRESS

m -7,077,000

IN -LB

+ 2,945,000

IN -LB

. i \-7,077,00Q

IN -LB

MOMENT

-21,400 psi + 44,900 psi

^+25,530 psi - 2,070 psi

J\ -21,400 psi

+44,900 psi COMBINED STRESS

AMPLITUDE, BENDING MOMENTS AND STRESSES

FOR TYPICAL H-SECTION MEMBER

FIGURE 6

473
120

no

£ 100

CL2— 90

UJ

_J5 SO

I

70

* 60

< 50 3

rft. Z

< 40

30

20

r-Vo^.AT BRIO 3E HEIGHT (240 FT. AVERAGE)

- v30 • A ' ct MFIfiUT ADnwr (lorn iwn

10 25 50

MEAN RECURRENCE INTERVAL - YEARS

100

240

20 25

PARTIAL ENLARGEMENT

HEIGHT ADJUSTMENT FORMULA

V240-(Wf* V30

V240 = 1.346 xV30

Source^ HCS Thorn, "New Distributions of Extreme Winds in the United

States", July 1968 Journal of the Structural Division, ASCE

EXTREME WINDS IN

THE AREA OF PHILADELPHIA

FIGURE 7

474
10 0.4 4 40

HOURS PER YEAR

200 400

Source: Merrison Committee Report, "inquiry Into The Basis of Design and

Method of

( Erection of Steel Box Girder Bridges , 1973. Data

taken for "Estuarial bridges or high valley crossings'.

DESIGN WIND DURATION

FIGURE s

475
ADDITIONAL BRACING , TYPICAL

27

ADDITIONAL BRACING ( same section as long, braces;

SUB-STRUTS

FOR

MINIMUM V r OF IOOM.RH.

SECTION A-A

FIGURE 9

476
ADDITIONAL 8RAC1NG TYPICAL

ADDITIONAL BRACING

PERMANENT CABLES

FOR

MINIMUM Vr OFIOOM.RH.

SECTION A-A

FIGURE 10

477
VARIATION IN ABSORBER EFFECTIVENESS

WITH MASS RATIO

FIGURE 11

478
FIGURE 12

479
VARIATION IN ABSORBER EFFECTIVENESS

WITH ABSORBER TUNING

FIGURE 13

480
4 .0--

I

o

o

3

Cl<

LU

Q.>

O

rr

3.0--

2.0--

.0--

9 = 0°^

Some coupling of weo

k

axis ond torsional motion

—o- No Absorbers A- With Absorbers

-0.87 I Modified mountma and added mertio)

Note: Absorbers tuned to optimum

frequency ratio.

-A: =A^-—i 1 1 1 20 40 60 80 100

PROTOTYPE WIND SPEED (MPH)

120

1:7.5 SCALE ELASTIC MODEL

WEAK AXIS MOTION VERSUS WIND SPEEDWITHOUT AND WITH TWO 0.005 SYNTHETIC

RUBBER ABSORBERS.

FIGURE 14

481
9 ,M 0 = 90°

No A bsorbers ---A- With Absorbers

2.0-

in \ f z \ 4^- = 0.70 ( Modified Mounting)

UNCHI

\

AMPLITUDE

i.o-

\j Note Absorbers tuned to

, optimum frequency rotio

PROTOTYPE

0 20 40 60 80 100 120 PROTOTYPE WIND SPEED (MPH)

STRONG AX 1 AND WITH

1-7.5 SCALE ELASTIC MODEL -

S MOTION VERSUS WIND S PEE D - W 1 T H OUT

TWO 0 005 SYNTHETIC RUBBER- ABSORBERS

FIGURE 15

482
30-

9=90°

CO

LJUJ

<ro

UJQ

LlJ Q

r>

20 --

•Motion limited to ovoid model damage —°— No Absorbers X-- Two 0005 Strong Axis Absorbers A Two 0005 Strong Axis and two 0 005 Weak Axis Absorbers

-r-=o.7,

t x

fz

= 0.87 ( Modified

mounting

)

UJ

OL>

o(-o

<r

a.

10 --

-x—i—*- 20 40 60 80

PROTOTYPE WIND SPEED (MPH)

I 00

U7.5 SCALE ELASTIC MODELTORSION AL MOTION VERSUS WIND SPEEDWITHOUT AND WITH SYNTHETIC RUBBER

ABSORBERS

FIGURE 16

483
VERTICAL ABSORBER

C_ Bracket

H-section web or box-section flange

Note: Absorbers will be mounted in pairs.

*7 for strong axis absorbers

8 0 bolt with hex head

and two nuts

Note: All plate

thicknesses ar

ro d unless noted.

ELEVATION

Drill S tap for cover box

screw (typ ) A

<L 3

8"0 bolt

Bent £. ^ ^~ z 0 hole

SECTION B-B

t"^"^ holes forV'0

high strength bolts SECTION A-A

Note: For box sections drill and tap

flange for 58"0 bolts Absorbers to be mounted on opposite sides of box in pairs.

ABSORBER DETAILS

FIGURE 17

484
CALCULATION OF THE GUST RESPONSES OF LONG-SPAN BRIDGES (II)

by

Nobuyuki Narita

Hiroshi Sato

ABSTRACT

The random responses of long-span bridges due to the turbulence of natural wind,

important in the fatigue problem of structural materials and in the serviceability of

public bridges are discussed. A simple formula by which the standard deviation of vertical

bending gust responses can be calculated is derived. It is proposed that designers of

long-span bridges can use this formula to assess gust responses, especially in the early

stages of design.

485
INTRODUCTION

In a previous paper [1] the strict method of calculating both the vertical bending and

the torsional mode of gust responses in the frequency domain and in the time domain was

described. Calculated results agreed reasonably with the observed ones, and it was found

that by setting more rational values for the parameters, the gust responses of long-span

bridges can be assessed more precisely by using the method. This strict method, however,

requires rather burdensome programming and computer operations. A simpler formula will be

helpful when designers of long-span bridges want to assess gust responses in round numbers

in the early stages of design. In this paper a simpler formula for calculating vertical

bending gust responses is proposed, and its derivation is explained.

NOTATION

B = Width of bridge deck

Cp = Lift coefficient defined for the structural axis

f = Frequency

fl = Natural frequency of the first vertical bending mode

Iw

= Intensity of turbulence of vertical component of wind

L = Span length of bridge

m = Mass per unit length of span

mg = Weight per unit length of span

R = Cross-correlation coefficient of wind

S = Power spectral density

U = Mean horizontal component of wind speed perpendicular to the bridge axis

u = Fluctuating horizontal component of wind speed perpendicular to the bridge axis

w = Fluctuating vertical component of wind speed

x = Spanwise coordinate of bridge

X = Aerodynamic admittance

a = Angle of attack

y = Constant for aerodynamic damping

Hi = Vertical bending displacement

i; s = Structural damping constant

486
p = Air density

a = Standard deviation

$1 = Vertical bending vibration mode of the first order

SIMPLE FORMULA FOR CALCULATING VERTICAL BENDING GUST RESPONSES

The standard deviation of vertical bending gust responses at the center of the main

span of a long-span bridge, whose first symmetrical vertical bending vibration mode

dominates in gust responses, can be calculated approximately with the following formula.

o

n (x) _ /0 = const • • I • U

2

• B 1/2

• f,

2

• L 1/3• (mg)

1 (1) x=L/2 da w 1

The proper value for the constant is found to be 0.03 - 0.06 when each value of

variables in the formula is substituted in the following unit.

o:cm, U; m/s, B;m, f , ; H7 , L; m, mg;

n Li. 1—m

DERIVATION OF THE FORMULA

When the above formula was derived, the factors which relate to quantities of vertical

bending gust responses were taken up first, the degree of influence of each factor was

investigated, then the shape of the formula was determined. Finally, the constant value

was calculated using the measured data of gust responses.

Holmes' method of calculating the gust responses was referenced when examining the

factors and investigating the degrees of influence. In Holmes' method, only the first

vertical bending vibration mode is considered, the same aerodynamic admittance and crosscorrelation coefficient is used for wind speed u and w, and the aerodynamic damping is

obtained from the quasi-steady theory. Thus, Holmes' method is simpler and less strict

than author's method [1], however, it can be said to be useful when it is referenced to

obtain this simple and approximate formula.

Holmes' calculation method is as follows.

,2

9 *,(x)

a

n (x) 2 =

(2irf

1

) 4 jy m$

1 (x) 2dx

SCfj) _i + /

1 S(f)df

c Jo

487

(2)
S(f) = (pBU)

2 X(f) 2

9 i dCp 2 C

F

S u(f) + Sw < f > 2 da

(3)

x [ [ * 1 (x) $

x (y) R(x, y, f)dxdy

•'O •'O

dc—F

^ pUB

? = C s + i§ T (4)

OTTmf

^

FACTORS IN DETERMINING THE QUANTITIES OF VERTICAL BENDING GUST RESPONSES

From Eq. (2) - Eq. (4), the factors in determining the quantities of vertical bending

gust responses of long-span bridges can be listed as follows.

Structure; vibration mode, natural frequency, mass per unit length of span, span

length of bridge, structural damping constant, width of bridge deck

Natural wind; air density, mean horizontal component of wind speed perpendicular to

the bridge axis, power spectral density of fluctuating component of wind speed, crosscorrelation coefficient of wind

Aerodynamic characteristics; aerodynamic admittance, lift coefficient and lift slope

defined for the structural axis, aerodynamic damping

DEGREE OF INFLUENCE OF EACH FACTOR

Most decks of long-span bridges have much smaller values for lift coefficients than

those of lift curves around angle of attack of zero degree. Using this fact, the following

proportional relationship can be derived approximately from Eq. (2) and Eq. (3).

a n (x) « • I w

• U

2

• r2 • m

_1 (5)

11 da 1

Next, a calculation was carried out using Eq. (2) - Eq. (4) in order to investigate

the degree of influence of factors whose influence cannot be obtained directly from Eq. (2)

and Eq. (3), and to recognize the relationship of Eq. (5). The fundamental parameters and

the calculated vertical bending gust responses using these parameters are shown in Table 1,

and Table 2, respectively. The gust responses were calculated changing one factor at a

time and are shown in Table 3. From Table 3, the degree of influence of each factor can

be assessed as follows.

488
Vibration mode; Two modes were considered. One was the mode of having the bridge deck

supported with hinges, and in the other mode supported rigidly. The difference of o

n between the two modes is less than 10 percent.

Natural frequency; <* fj2 '^

_n Q A Q Mass per unit length of span; « m

-0 3 Span length of bridge; « L

Structural damping constant; The difference between with c s of 0.003 and a n with £ s

of 0.03 is about 10 percent even where the aerodynamic damping has as small a value as

possible for the ordinary deck of long-span bridges.

Width of bridge deck; a <* B 0,6

Air density; This can be neglected because its variation seems to be very small.

Mean horizontal component of wind speed perpendicular to the bridge axis;

n Power spectral density of wind speed fluctuating component; Instead of considering the

influence of P.S.D. directly, the influence of intensity of turbulence and that of the distribution of power in the frequency domain were considered separately. As for intensity of

turbulence, a « I , when I and I change their values with a constant ratio. In order ' n. w ' u w

°

to examine the influence of distribution, three formulae of Sw

(f) by Panofsky and

McCormick, Busch and Panofsky, and Singer, Busch and Frizzola [3] were considered. In

doing this, their influence on was found to be about 10 percent.

Cross-correlation coefficient of wind and aerodynamic admittance; It is difficult to

set the precise values of these two factors, and unfortunately their effect, especially of

the aerodynamic admittance, cannot be neglected. Therefore, these effects were considered

in determining the constant value in Eq. (1) from measured data.

0.8-0.9

r dCF , Lift slope for the structural axis; a = {

J

1 da

Aerodynamic damping; In comparing unsteady aerodynamic coefficients of a bridge deck

measured in a wind tunnel tests with Eq. (4) derived from the quasi-steady theory, almost

all the experimental coefficients fall between y of 0*5 - 1.0. When y is varied from 0.5

to 1.0 with c, s of 0.003, an varies about 20 percent.

489
By using the above indicies in round numbers and neglecting factors of small

influence, the shape of the formula can be determined as in Eq. (1).

DETERMINATION OF THE CONSTANT VALUE

The constant value in Eq. (1) was calculated using the data shown in Table 4. The

calculated constant values - using the data measured at the sites of the Kammon Bridge,

Suehiro Bridge, Hirado Bridge, and in the wind tunnel test of the West Gate Bridge [2] ~

were 0.047, 0.032, 0.039 and 0.026 - 0.059, respectively. Thus, the proper value for the

constant can be said to be 0.03 - 0.06. In the above calculation, the data from the Kammon

Bridge and the Suehiro Bridge were used in a form of an approximated equation as derived

2 from the data with I at about 0.10 under the assumption that a„ <* U . The data used from w n the Hirado Bridge have the largest U value in any of the data obtained so far.

REFERENCES

[1] Narita, N. and Sato, H. , "Calculation of the Gust Responses of Long-Span Bridges,"

11th Joint Meeting, Japan-U.S. Panel on Wind and Seismic Effects, UJNR, Tsukuba,

September 4-7, 1979.

[2] Holmes, J. D. , "Prediction of the Response of a Cable Stayed Bridge to Turbulence,"

Proc. 4th International Conference on Wind Effects on Buildings and Sites, 1975.

[3] Okauchi, I., Ito, M. , and Miyata, T. , "Structures Resistant to Wind," 1977 (in

Japanese)

.

490
Table 1. Fundamental Parameters

Vibration Mode

Natural Frequency

Weight per Unit Length of Span

Span Length of Bridge

Structural Damping Constant

Width of Bridge Deck

Mean Horizontal Component of Wind

Speed Perpendicular to the

Bridge Axis

Power Spectral Density of Fluctuating Component of Wind Speed

Decay Factor of Cross-Correlation

Coefficient of Wind

Aerodynamic Admittance

Lift Slope Defined for the Structural Axis

Aerodynamic Damping

Table 2. Calculated Vertical Bending

U=18.5m/s

37.0

49.3

calculated value for West Gate Bridge

0.33Hz

20.10 ton/m

336m

0.013

37.4m

18.5m/s, 37.0m/s, 49.3m/s

Su; due to Harris

Sw ; due to Busch & Panofsky

I u ; 0.20 Iw ; 0.10

7.0

due to Holmes suggested by Vickery

5.10

quasi-steady theory with constant of

0.60

Gust Responses with Parameters in Table

0, (L/2)

0.028cm

0.14

0.27

491
Table 3. Degree of Influence of Each Factor (Calculation Result)

Vibration Mode (L=848m)

°r, (L/2)(cm)

Support Hinge Rigid

U=18 5m/s 0.025 0.027

37 0 0.11 0.12

49 3 0.21 0.23

Natural Frequency f^

°,, (L/2)(cm)

^(Hz) 0.0825 0.165 0.330 0.660

U=18. 5m/s 0. 65 0.14 0.028 0.006

37.0 2.4 0.65 0.14 0.03

49.3 3.7 1.1 0.27 0.05

Weight per Unit Length of Span mg

mg(ton/m) 5.0

ML/2) (cm)

10.0 20.

1

40.0

U=18.5m/s 0.099 0.053 0.028 0.015

37.0 0.43 0. 24 0.14 0.08

49.3 0. 79 0.46 0.27 0.16

Span Length of Bridge L (Mode for Hinge-Support)

L(m) 212

o, (L/2) (cm)

424 848 1696 3392

U=18.5m/s 0.037 0.030 0.025 0.021 0.019

Structural Damping Constant C,

0, (L/2) (cm)

0.0032 0.0064 0.013 0.025

U=18. 5m/s 0.031 0.030 0.028 0.027

37.0 0.15 0.15 0.14 0.13

49.3 0.29 0.28 0.27 0.25

Width of Bridge Deck B

B(m) 12

0, (L/2) (cm) 24 37.4 48

U-18.5 0.014 0.021 0.028 0.034

37.0 0.070 0.11 0.14 0.16

49.3 0.14 0.21 0.27 0.31

492
Table 3. (Continued)

Intensity of Turbulence Iw (I u=Iwx2.0)

°r (L/2)(cm)

lw 0.025 0.05 0 10 0.15 0.20

U=18.5m/s 0.007 0.014 0 028 0.042 0.056

37.0 0.04 0.07 0 14 0.21 0.28

49.3 0.07 0.14 0 27 0.41 0.54

Shape of Power Spectral Density ^

s

0, (L/2) (cm)

Panofsky & Busch & Panofsky Busch & McCormick Frizzola

U=18.5m/s 0.028 0. 028 0.027

37.0 0.13 0. 14 0.13

49.3 0.26 0. 27 0.24

Decay Factor of Cross-Correlation Coefficient of Wind k

k

0, (L/2) (cm)

1.0 3.5 7 0 14.0 28.0

U=18.5m/s 0.047 0.034 0 028 0.023 0.020

37.0 0.24 0.18 0. 14 0.11 0.09

49. 3 0.43 0.34 0 27 0.21 0.17

Aerodynamic Admittance X

X

Or (L/2) (cm)

Liepmann Holmes Davenport (k=7 . 0)

U=18. 5m/s 0.020 0. 028 0.024

37.0 0.09 0 14. 0.12

49. 3 0.18 0. 27 0.24

Lift Slope Defined for the Structural axis

dCF

dCf

0, (L/2) (cm)

1.0 2.0 3 0 4.0 5.1 6.0

U=18. 5m/s 0.007 0.012 0 018 0.023 0.028 0.033

Constant in Aerodynamic Damping r

7

o r, (L/2) (cm)

(^=0.0032) 0.5 1 0 (C^O.0047) 0.5 1.0

U=18.5m/s 0.032 0 028 0.031 0.028

37.0 0.16 0 13 0.16 0.13

49.3 0.31 0 .25 0.30 0.25

493
Table 4. Data Used for Determination of the Constant Value in Eq. (1)

Kammon Br

.

Suehiro Br. Hirado Br. West Gate

1

Br . **

2 3 dCF /do 3.55 2.86 3.31 5.10 5.10 5.10

0.10 0.10 0.10* 0.10 0.10 0.06

U (m/s) *** *** 18.5 18.5 37.0 18.5

B (m) 29.0 18.5 14.5 37.4 37.4 37.4

fl(Hz) 0.20 0.48 0.31 0.33 0.33 0.33

L (m) 712 250 460 336 336 336

mg (ton/m) 24. 7 9. 26 11.0 20.1 20.1 20.1

°i (L/2) l.OxlO

-2

3.0xlO" 3 2.0 2.8 16.6 1.1

(cm) xU

2 xU

2

const

.

0.047 0.032 0.039 0.040 0.059 0.026

est imated value

** Three experimental data were used which have maximum (2),

2 minimum (.3), and intermediate (1) values of ° Tl /U .

2 *** These data were used in the form of 0, = const. xU .

494
ACTIVE TURBULENCE GENERATOR FOR AERODYNAMIC

TESTS OF BRIDGE SECTION MODELS

by

B. Bienkiewicz

J. E. Cermak

J. A. Peterka

INTRODUCTION

Turbulence in an oncoming flow has profound effects on aerodynamic behavior of

suspension bridges. Even simple wind-tunnel tests with a typical section of a bridge deck

(the so-called bridge section model) indicate — at least qualitatively — the main changes

introduced by turbulence. In a typical situation turbulence usually transfers occurrence

of flutter-type aerodynamic instability to higher values of the reduced velocity. At the

same time random oscillations of a bridge deck (buffeting) — not present in a smooth

flow — appear at lower values of reduced velocity. Their magnitude ordinarily increases

with the wind speed, Cermak et al. [1].

Wind-tunnel tests with bridge section models are less expensive than wind-tunnel tests

with full bridge models. Also they can be conducted In smaller wind tunnels. Typically,

turbulence for a section-model test is introduced into the flow by placing a grid of rectangular or cylindrical rods upstream of the test section. The main characteristics of

turbulence generated in such a manner, especially integral length scale, depend on the

wind-tunnel size. If a wind-tunnel with a relatively small test section is considered,

the attainable integral length scale is usually smaller than the width of a bridge deck

scaled for the tunnel. On the other hand, full-scale measurements show that the integral

length scale of the atmospheric turbulence is several times longer than the width of a

typical bridge deck, Scanlan [2], It has not been determined yet what the effects of the

integral length scale of turbulence are on a bridge-deck aerodynamic response. In order to

explore this question in more detail it is necessary to conduct a series of tests with

bridge section models where the turbulence integral length scale could be easily changed to

attain values far beyond the limits imposed by grid turbulence.

495
This paper describes a concept and results of initial testing of a turbulence

generator which attempts to provide the desired flexibility for control of the integral

length scale.

ACTIVE TURBULENCE GENERATOR

The concept for an active turbulence generator was described earlier in some detail by

Cermak et al. [3]. It is proposed that a generator of the same type be constructed in the

George S. Vincent Wind Tunnel shown in Figure 1. The main components of the generator are

two arrays of symmetrical airfoils (A and C) oscillating 180 degrees out of phase and in

phase, respectively. The array A is assumed to produce primarily longitudinal velocity

fluctuations whereas array C is intended to generate primarily fluctuations of transverse

velocity component. The driving mechanism of the generator contains components of a standard electro-hydraulic servo-system. The system is driven by input signals recorded on an

analog magnetic tape. It is expected that in order to improve "control" of the two generated turbulent velocity components — longitudinal and transverse — some flow suppressors

B shown in Figure 1 will be necessary.

The concept has been initially tested in the Fluid Dynamics and Diffusion Laboratory,

Colorado State University, as a part of a study sponsored by the Federal Highway Administation, Department of Transportation (FHWA, DOT). The generator was placed in the opencircuit wind tunnel as shown in Figure 2. The two arrays of symmetrical airfoils NACA 0015

made of extruded aluminum were driven by hydraulic actuators controlled by a MTS Corporation servo-system. Two kinds of input driving signals were used. Initially an analog

signal — Gaussian white noise with low-pass filtering — produced by a standard noise generator (Hewlett-Packard) was used to drive the arrays. Later a computer-simulated signal

after digital-to-analog conversion was fed as input to the driving system. During a Monte

Carlo computer simulation the time series of a given spectra was generated using procedure

suggested by Shinozuka (4) and Wittig et al. [5]. Use of a FFT "stacked" routine, Hudspeth

et al. [6] , resulted in a substantial reduction of computer time required for simulation.

For most of the experiments only one array of oscillating airfoils was activated and

the resulting turbulent velocity field downstream of the generator was evaluated. Some

results obtained with the array of airfoils oscillating in phase will be presented in the

next paragraph.

496
EXEMPLARY RESULTS

During initial tests with the active turbulence generator a driving input signal was

obtained from an analog noise generator. An example of such a signal — Gaussian white

noise with a low pass filtering — is shown in Figure 3. It can be seen from this spectrum that signal was filtered with a frequency cut-off equal to approximately 5 Hz. This

signal was used to drive the generator. Spectra of the turbulent velocity components were

taken downstream of the generator. Figure 4 shows such spectra for the case when the generator contains only one array of airfoils oscillating in phase (vertical mode). The generated turbulence is approximately isotropic (I v

= 6.5 percent = I u

= 7.6 percent and Lu

=

Lv ) , Batchelor (7), and spectra can be fitted by the von Karman formulas — smooth

curves —, von Karman [8]. The resulting integral scale of turbulence is relatively low

(0.17 W where W is width of the wind-tunnel test section); however, it is larger than scales

obtained using a typical grid to generate turbulence.

In the next step, a computer simulated signal of a given spectra was used to drive the

generator. A typical spectrum of a simulated signal compared with a target spectrum is

shown in Figure 5. The digital signal and its analog counterpart are presented in

Figure 6. It can be seen that all the features of the digital signal are preserved during

the digital-to-analog conversion. As a target spectrum for the vertical velocity component

an expression proposed by Panofsky et al. [9] was chosen. The frequency cut-off of the

simulated signal shown in Figure 6 was 20 Hz. The same signal was used to determine a

transfer function of a control-driving system. As can be seen in Figure 7 the system has a

good frequency response up to roughly 15 Hz.

Spectra of vertical velocity component generated using the mentioned signal (see

Figures 5 and 6) and the array of airfoils oscillating in phase (vertical mode) is shown in

Figure 8. The spectrum obtained and the target spectrum match quite well except for a

spike at a reduced frequency nz/v of about 2. The corresponding integral length scale was

about 1.2 ft which indicates a substantial improvement when compared with scales of

turbulence generated using analog Gaussian white noise with a low-pass filtering. In

order to investigate the origin of the spike in Figure 8 a series of tests was conducted.

A computer simulated signal with lower frequency cut-off, shown in Figure 9, was used in

turbulence generation. For this case, see Figure 10, the agreement between the generated

497
and target spectra of vertical velocity component was worse but at the same time the spike

was less pronounced. The integral scale of turbulence was found to be 2.5 ft.

Finally the origin of the spike in the turbulent velocity spectra was fully explained.

Measurements of wind-tunnel oscillations showed that the frequency at which the spike

occurred coincided with the lowest natural frequency of the wind tunnel, compare Figures 8

and 11. Even small oscillations of the wind tunnel, excited by supports of the oscillating

airfoils attached to the tunnel, developed pronounced effects on the turbulence spectra as

was shown by the data presented.

REMARKS

The results obtained to date indicate that the turbulence integral length scale can be

controlled by the input driving signal. The ratios of integral length scale to modelbridge width that appears to be realizable are similar in magnitude to those for full-scale

bridges in natural winds. Even small oscillations of a wind-tunnel test section that

encloses the flow (and generator) can introduce large disturbances in a turbulent velocity

field developed by an active generator. Therefore, the driving system of the generator

and the generator itself should be isolated from the wind tunnel. Further measurements of

spectra should be made when this has been accomplished for the present system. However,

this phenomena should not be a problem for active turbulence generation in an open-jet

wind tunnel such as the George S. Vincent wind tunnel shown in Figure 1.

ACKNOWLEDGMENTS

Support for the study described in this paper was provided by FHWA, DOT through

Contract No. DOT-FH-1 1-9233.

498
REFERENCES

Cermak, J. E. , Bienkiewicz, B. , and Peterka, J. A., "Effects of Turbulence on Aerodynamic Characteristics of Bluff Bodies and Suspension-Bridge Decks," presented

at 1978 FCP Research Review Conference, College Park, Maryland, October 1978.

Scanlan, R. H. , "Recent Methods in the Application of Test Results to the Wind Design

of Long, Suspended-Span Bridges," Report FHWA TD-75-HS, Federal Highway Administration, Washington, D.C., October 1975.

Cermak, J. E. , Bienkiewicz, B. , Peterka, J. A., and Scanlan, R. H. , "Active

Turbulence Generator for Study of Bridge Aerodynamics," Proceedings Third Engineering

Mechanics Division Specialty Conference , September 17-19, 1979, The University of

Texas at Austin, Austin, Texas, pp. 183-186.

Shinozuka, M. , "Digital Simulation of Random Processes in Engineering Mechanics with

the Air of FFT Technique," in Stochastic Problems in Mechanics , Waterloo Press,

Waterloo, Ontario, Canada, 1974, pp. 277-286.

Wittig, L. E., and Sinka, A. K. , "Simulating of Multicorrelated Random Processes

Using FFT Algorithm," Jl. Acoust. Soc. of America , Vol. 58, No. 3, September 1975,

pp. 630-634.

Hudspeth, R. T. , and Borgman, L. E., "Efficient FFT Simulation of Digital Time

Sequences," Jl. Engrg. Mech. Div . , ASCE, Vol. 105, No. EM2 , April 1979, pp. 223-234.

Batchelor, G. K. , "The Theory of Homogeneous Turbulence," Cambridge at the University

Press, 1956.

von Kafmari, T., "Progress in the Statistical Theory of Turbulence," Proc. Nat. Acad. Sci. , Vol. 34, 1948, pp. 530-539.

Panofsky, H. A., and McCormick, R. A., "The Spectrum of Vertical Velocity Near the

Surface," Quart. Jl. , Royal Meteorological Soc , 1960, Vol. 86, pp. 495-503.

499
Figure 1. George S. Vincent Wind Tunnel, FHWA Fairbank Research Laboratory, DOT

Figure 2. Structural Wind Tunnel

500
10

Figure 3. Analog Input Signal

FREQUENCY, Hz

501
502
Figure 6. Digital and Analog Input Signal for Vertical Mode

COMPUTER SIMULATION

VERT VEL COMP CPANOFSKY)

1 1

TIME - SEC

Figure 7. Transfer Function for Mechanical Driving Signal

FREQUENCY - HZ

H 3.80$'

5 80 18 00 15 80 28,< VERTICAL MODE - PSEUDORANDOM SIGNAL

503
Figure 9. Spectra of Simulated Input

Signal with Low Frequency Cut-off

COMPUTER SIMULATION

VERT VEL CCMP/CPAI\; C~SK Y :

:j«z/j

504
0)

TURBULENCE GENERATED

VERT VEL COMP CPftNOFSA'

Figure 10. Generated Turbulence Using

Signal with Low Frequency Cut-off

N*Z/U

Figure 11. Spectra of Wind Tunnel Oscillations

* S

*

</) oo z -

a: s \

WIND TUNNEl oscIlL

VERT MODE

305
Report of the Task Committee on

(A) Strong-Motion Instrumental Arrays and Data

UJNR PANEL ON WIND AND SEISMIC EFFECTS

Date : May 20, 1980

Place : National Bureau of Standards

Washington, D.C. U.S.A.

Task. C r oup : U.S. Side - Gutam Bagchi (Acting Chairman)

Japan Side - Hajime Tsuchida (Chairman)

(1) Activities and Principal Accomplishments to Date

Catalogues of strong-motion earthquake records observed both in

the United States and Japan are being exchanged. U.S. data are

published in the "Seismic Engineering Program Report," and the

Japanese data are reported in the publication, "Strong-Motion

Earthquake Records in Japan."

Ihe Imperial Valley Earthquake of October 15, 1979, and the

Coyote Lake Earthquake of August 6, 1979, were discussed in a paper

and reports containing a catalogue of the event records were provided

for the Japanese side. One of the Japanese papers described their

Dense Instrumentation Array Program. This program promises to provide

the most elaborate data necessary to enhance the state-of-knowledge

in ground-motion propagation.

(2) Future Programs

The both sides of the Task Committee plan to continue their

respective efforts in the following eight points:
1) After an earthquake which has caused damage to structures or an

earthquake during which maximum acceleration exceeding about 0.1 G

has been recorded, the task committee of a country where the

earthquake has taken place will provide a list of the strong-motion

earthquake records for the counterpart of the task committee. The

list contains maximum component acceleration of each record. If

there is such a list compiled by any organization, the list mentioned

above may be replaced by it.

2) Every year the task committee will exchange catalogs of the

strong-motion earthquake records in thelast year. 'The ;

catalog contains maximum component accelerations and wave forms

of major records. If there is such a catalog compiled by any

organization, the catalog mentioned above may be replaced by it.

3) The task committee makes appropriate arrangements to provide

digitizable copies of records when they are requested. In addition,

arrangements will be made to provide information on the

characteristics of the site and structures at the locations where

such records are obtained.

A) When the organizations taking part in the panel publish reports

on the strong-motion earthquake records, the organizations will

distribute copies of the reports to the organizations of the

panel interested in them. The task committee exchanges lists

of the organizations which wish to receive the reports.

5) Every year at the time of the joint meeting the task committee

exchanges lists of reports on the strong-motion earthquake

records and on analysis of the records published in the last year.

The list will cover all the reports on the above mentioned topics

published by the organizations taking part in the panels and as many

reports as possible published by other organizations and universities.

-m>7
The task committee will continue to exchange lists of

digitized data on all the major strong-motion earthquake

records recovered in both countries. Besides the digitized

records in form of lists, an exchange of digitized records in

form of computer magnetic tapes, cards, and other media will

be continued.

The task committee plans to assist and cooperate, where possible,

in the following areas.

a) Assistance and cooperation with governmental organizations in

other seismic areas, in order to promote high quality strong

earthquake motion observations in all seismically active areas

of the world.

b) Assistance and cooperation with any international effort to

record strong ground motion close to the source of a large

magnitude shock.

The task committee feels the UJNR Panel on Wind and Seismic Effects

should play a major cooperative role in the implementation of relevant

parts of the Resolution of the International Workshop on Strong-Motion

Earthquake Instrument Arrays held May 2-5, 1978, in Honolulu, Hawaii,

USA, and particularly support the recommendation to install dense

arrays of strong-motion instrument (See Appendix I). The exchange

of complete information on all aspects of the program, as it develops

particularly in Japan and the United States, will be carried out in

the manner of our standard exchange when appropriate.

508
-Appendix I Resolution of the International Workshop

on Strong-Motion Earthquake

Instrument Arrays

(May, 1978, Honolulu, Hawaii, U. S. A.)

The following resolution was approved unanimously by. the delegates

of the International Workshop on Strong-Potion Earthquake Instrument

Arrays during a general session on Kay 5, 1978.

The protection of life and property from the devastating effects

of earthquakes is an urgent world-wide problem. An understanding of the nature of strong earthquake nations is of crucial importance in solving

this problem. At the present time, however, there is a scarcity of

engineering data acquired near the centers of destructive earthquakes,

and existing instrument arrays are inadequate to provide the necessary

data. Yet there is a high probability of occurrence of destructive

earthquakes in different parts of the world in the next decade. The

participants in this international workshop unanimously recommend that

the earthquake-threatened countries and other concerned countries and

organizations make a concerted effort to establish a comprehensive

world-wide system of specialized strong-motion earthquake instrument

arrays capable of resolving the nature of the earthquake source

tachanism, wave propagation and local site effects. As a first step,

the following specific recommendations should be implemented.

1. The International Association for Earthquake Engineering in collaboration with the International Association of

Seismology and Physics of the Earth's Interior form an

International Strong Motion Arrays Council to facilitate

the establishment of strong-ration earthquake instrument

arrays.

Earthquake-threatened countries individually and collectively initiate the immediate installation of minimal

arrays of 10-20 strong-motion instruments at least at

the 28 world-wide sites identified by this workshop.

High priority be given to the design and installation of

roore elaborate source mechanism, wave propagation and

local effects arrays, particularly at the six critical

sites Identified.

A mobile strong-motion instrument array capable of making

source mechanism, wave propagation and local effects

measurements be established and maintained for deployment

immediately following the occurrence of a major earthquake

for. the recording of aftershocks.

509
Report of Task Committee on

(B) Large-Scale Testing Program

UJNR PANEL ON WIND AND SEISMIC EFFECTS

Date : May J 9 , 1980

Place : National Bureau of Standards

Gaithersburg, Maryland U.S.A.

Attendees : Japan side - Keiichi Ohtani (Co-Chairman)

Makoto Watabe

U. S. side - Edgar V. Leyendecker (Co-Chairman)

Hai Sang Lew

(1) Activities and Principal Accomplishments to Date

The status of the U.S. -Japan Joint Earthquake Research Program Utilizing LargeScale Testing was discussed, and for the purpose of the Panel's endorsement the

available U.S. -Japan Joint Research Program Plans were reviewed.

(2) Future Programs

The Task Committee adopted the following resolutions as a result of discussion:

A. The Co-Chairmen of both sides of the UJNR Panel on Wind and Seismic Effects

have received a request for endorsement of the U.S. -Japan Joint Earthquake

Research Program Utilizing Large-Scale Testing in a letter signed jointly by the

Science and Technology Agency and the Ministry of Construction of Japan and the

National Science Foundation of the U.S., referred to herein after as the Parties.

The Panel has previously expressed its support for the Large-Scale Testing Program.

B. This Task Committee recommends that the program and its components proposed

for endorsement by the Panel be submitted to the Task Committee for its review

and comments. Based on this review, the Committee Co-Chairmen will prepare

recommendations to the Panel Co-Chairmen at as early a date as possible. The

Task Committee recommends that the Panel Co-Chairmen jointly notify the

requesting Parties of those proposals receiving endorsement, along with Committee

review comments.

C. The Task Committee recommends that close liaison be maintained between the

Panel and the Patties. This should include notification of proposed research

direction and change in direction and should also include a periodic review of

progress of the joint research program. The periodic review should include

presentations to the Panel at one or more of its joint meetings.

D. The Task Committee will review progress of the Joint Research Program and

continue to reassess the need for testing of other structures and report its

findings and recommendations to the Panel.

E. Membership of the Task Committee should be expanded to include larger and more diverse membership. In particular, representation from (F) Disaster Prevention

Methods for Lifeline Systems and (C) Repair and Retrofit of Existing Structures

Task Committees should be included.

510
REPORT OF TASK COMMITTEE ON

(C) REPAIR AND RETROFIT OF EXISTING STRUCTURES

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS

May 19, 1980

Gaithersburg, Maryland

PRESENT: M. Watabe Japan

R.D. McConnell USA

R.J. Morony USA

J.D. Cooper USA

J.B. Scalzi USA

ACCOMPLISHMENTS TO DATE - DURING 1979-1980 :

A grant was made to Dr. Robert D. Hanson, University of Michigan to serve

as Project Manager for the U.S. Task Committee.

By correspondence with Dr. Kiyoshi Nakano, he established areas of interest to be the repair and retrofit of buildings and bridges. Buildings is interpreted

as general structures.

At the 11th Joint Meeting of UJNR it was decided to hold a one day seminar

and one day field trip preceding the 12th Joint Meeting. The seminar to be held

in L.A. As a result, a meeting of U.S. experts was held in December of 1979

in Washington to discuss topics to be presented and speakers to be invited

from the U.S. side. Correspondence with Japan listed the topics and requested

speakers

.

The seminar was held in Los Angeles at the offices of Dr. M. Agbabian.

Many interesting papers were presented by both sides. The proceedings will

be published by Dr. Hanson and distributed to the members and others. The

field trip included a visit to bridge sites to view the cable restrainers which

were being installed. Three bridges were inspected. A visit to a school

(John Marshall) which was being restored for preservation was very interesting.

The local residents wanted to preserve the school because of its architectural

beauty. The cost of the repairs and new installations were a little less than

building a new school.

Discussion at Los Angeles among the members present included:

1. Task Committee to confine itself to structures

2. Lifeline systems was to be include at a later date. However, a member

(preferably chairman) of the ASCE TCLEE Committee is to be invited

to participate.

511
3. Next meeting to be in Japan. Exact time and place to be determined

by co-chairmen.

4. Research topics are to be studied and presented as areas of cooperative

effort in the future.

MEETING OF MAY 1980 : Progress to date was reviewed and discussed. Topics discussed were:

1. Format (title page of seminar proceedings) to include UJNR Task

Commi ttee.

2. It is suggested that Dr. Hanson contact other people for papers

to be included in the proceedings in particular James Lefter of the

Veterans Administration.

3. Seminar proceedings to be distributee to all agencies, practitioners,

reserachers, and NTIS.

4. Papers in native language to be translated by each side for

distribution in their country.

5. Recommended that scope of Task Committee activities be reviewed in

light of interest by members of Lifelines Committee.

6. Recommend that activities from U.S.- U.S.S.R. and U.S. -PRC groups be

reported at next meeting.

7. Recommend that topics of economics and cost-effectiveness be delayed

until technical problems have been reasonably solved.

8. Suggest that next meeting of the Task Committee be held in conjunction

with UJNR meeting in May 1981, or at time of performinq large scale

test of reinforced concrete building.

9. Recommend that exchange of data be a continuous activity between

members as well as co-chairmen.

10. Recommend that exchange of data include non-structural components,

elevators, parapets, etc.

11. Recommend that discussion of single family residences be included

in next seminar.

512
Report of Task Committee on

(D) Evaluation of Performance of Structures

UJNR PANEL ON WIND AND SEISMIC EFFECTS

Date : May 20, 1980

Place : National Bureau of Standards

Gaithersburg, Maryland U.S.A.

Attendees : U.S. Side - G. Robert Fuller (Co-Chairman)

John B. Scalzi

Richard McConnell

Japan Side - Makoto Watabe (Co-Chairman)

I. Progress 1979-1980

A considerable amount of effort has been expended in both the

U.S. and Japan to develop methodologies and procedures for evaluating

the seismic resistance of existing civil engineering structures.

Closely related are the projects involving repair, retrofit, strengthenin

and rehabilitation of structures.

A. U.S. Activities :

1 . Federal Interagency Committee on Seismic Safety in Construction

(ICSSC) : This organization is made up of representatives of

all U.S. Federal Government agencies involved in construction

of buildings and lifeline structures. The Earthquake Hazards

Reduction Act of 1977 charges the President with establishing

and maintaining an effective earthquake hazards reduction

program. ICSSC was originally organized by the Presidents'

Office of Science and Technology Policy, but now is under the

auspices of the Federal Emergency Management Agency (FEMA)

.

513
Ten subcommittees, under ICSSC, were established to develop

seismic design and construction standards for Federal buildings,

rehabilitation of existing buildings, and also buildings

constructed under grant, loan, leasing and regulatory programs.

Subcommittee Three - Existing Buildings is establishing

procedures for evaluating the seismic resistance and for

strengthening existing buildings. This criteria will be

completed later in 1980 for review and testing by Federal

agencies, prior to adoption and implementation in 1981.

Building Seismic Safety Council (BSSC) : This Council was

established by private industry, under the auspices of the

National Institute of Building Sciences, to develop aseismic

design and construction recommendations to be adopted by model

building codes, local municipalities, and the design profession.

BSSC established nine technical committees to review and refine

the Applied Technology Council document ATC 3-06, "Tentative

Provisions for Development of Seismic Regulations for Buildings.'*

Technical Committee Nine (TC-9) on Regulatory Use is charged

with reviewing ATC 3, Chapters 1 - "General Provisions" and

Chapter 13 - "Systematic Abatement of Seismic Hazards in Existing

Buildings." Chapter 13 contains provisions for identifying

buildings requiring evaluation, evaluation procedures (qualitative

and analytical) and hazard abatement measures.

Chapter 14 of ATC 3-06 contains guidelines for repair and

strengthening of existing buildings which includes a "Checklist

for Existing Building Data File."

M4
Chapter 15 provides "Guidelines for Emergency PostEarthquake Inspection and Evaluation of Earthquake Damage in

Buildings." This includes:

a) Selection of Inspection Personnel

b) Training Programs

c) Inspection Equipment

d) Procedures for Inspection

e) Evaluation of Structural Damage

f) Evaluation of Nonstructural Damage

g) Evaluation of Auxiliary Systems

h) On-Site Soil and Foundation Conditions

1) Tsunami and Seiche Effects

j) Reinspection, Repairs, and Records.

Housing and Urban Development (HUD) : The "Methodology for

Seismic Evaluation of Existing Multistory Residential Buildings"

was published in November 1978, and reported by Mr. Fuller at

the 10th Joint Panel Meeting held in May 1978. A presentation

was then made by Mr. Fuller at a Workship on Earthquake-Resistant

Repair and Retrofit of Buildings, held on May 16, 1980 in

Los Angeles, California.

This "Seismic Evaluation and Rehabilitation of HUD

Residential Buildings" paper included the two example buildings

in the Methodology plus several other buildings evaluated and

strengthened using the methodology. Several structures in

Los Angeles, San Francisco, Oakland, Memphis (TN) , Rochester (NY),

and St. Louis (MO) have been reviewed, analyzed and strengthened.

515
B. Japan Activities ;

1. Over all investigations on reinforced concrete building

structures have been conducted, especially in Tokija) Districts.

2. Repair or retrofitting of steel structures has been accomplished

in Sendai area.

3. An automatic computer evaluation system for reinforced concrete

building structures has been completed.

4. A practical procedure to calculate ultimate shear strength of

building structures in accordance with the promulgation of

"New Aseismlc Enforcement of Buildings," has been developed.

5. A simple aseismic capacity evaluation procedure for wood-framed

houses has been drafted by the Building Disaster Prevention

Association and the Ministry of Construction.

Resolutions

Because of the recent increase in the use of various evaluation

and analysis techniques for assessing the earthquake resistance of

existing structures, an attempt should be made to exchange reports of

projects which have been evaluated and strengthened.

A. Establish a bibliography of evaluation methodologies and analysis

techniques in cooperation with the Task Committee on Repair and

Retrofit.

B. Maintain a file in each country and exchange technical data on

projects analyzed for seismic resistance and strengthened.

C. Develop recommendations for evaluating structures to assess

earthquake hazard potential. Include recommendations for postearthquake evaluation of performance and damage assessment of

structures

.

516
D. Instrument existing structures that have been evaluated and

strengthened and monitor performance during earthquakes. Conduct

post-earthquake investigations and damage assessment of these

structures

.

517
Report of the Task Committee on

(E) Land Use Programs for Controlling Natural Hazard Effects

UJNR PANEL ON WIND AND SEISMIC EFFECTS

Date : May 21, 1980

Place : National Bureau of Standards

Gaithersburg, Maryland U.S.A.

Attendees : U.S. side - G. Robert Fuller (Chairman, Temporary)

Japan side - Hajime Tsuchida (Chairman, Temporary)

A. Activities

1. It was reported that in January 1980 Mr. T. Tazaki, Public Works

Research Institute, Ministry of Construction, completed his study

at UCLA on socio-economic aspects of land use programs, in

cooperation with U.S. researchers.

B. Resolutions

1. The resolutions of this Task Committee from the 8th through 11th

Joint Panel Meetings are still in effect and will continue.

2. The Task Committee recognizes that methodology on the land use programs should be urgently developed considering the implementation

of the Earthquake Hazards Reduction Act of 1977 (Public Law 95-124)

in the U.S.A. and the Large Scale Earthquake Countenneasures Act,

Law No. 73 (1978.6.7) in Japan.

3. The exchange of papers and references on land use planning should

be continued, especially in the following areas:

a) Current methods of earthquake hazard risk analysis,

b) Applications of the methods contained in the above

earthquake hazard mitigation acts, and

c) Hazard maps for earthquake disaster prevention planning,

including socio-economic considerations.

4. Cooperation and exchange of engineers, geophysiclsts , land use planners and social scientists should be encourgaed in order to promote more effective land use planning.

518
Report of the Task Committee

on

(F) Diaster Prevention Methods for Lifeline Systems

UJNR PANEL ON WIND AND SEISMIC EFFECTS

Date : May 20, 1980

Place: National Bureau of Standards, Gaithersburg , Maryland, U.S.A.

Attendees : Japan side - Keiichi Ohtani (Chairman, Temporary)

National Research Center for Diaster Prevention

Science and Technology Agency

Yasuyuki Koga

Public Works Research Institute

Ministry of Construction

U.S. side - Charles F. Scheffey (Chairman)

Federal Highway Administration

James D. Cooper

Federal Highway Administration

Haaren A. Miklofsky

Federal Highway Administration

Teoman Ariman

Dept. of Aerospace and Mechanical Engineering

University of Notre Dame

(1) Activities and Principal Accomplishments to Date

Through the exchange of letters between the chairmen of both sides, an initiation of immediate planning, for use of the large-scale testing facilities

of both countries to resolve critical problems in the behavior of lifeline

systems, has been undertaken.

From the U.S. side, several professors concerned visited Tsukuba Science

City and exchanged information on earthquake engineering of lifeline systems and

view of cooperative research program.

(2) (Description of discussions carried out the meeting.)

(3) Future Programs

The Task Committee adopted the following resolutions as a result of

discussion:

1) The present area and level of information exchange on damage studies,

research information, design standards criteria, and administrative regulations

with regard to lifeline safety should be continued, with expanded effort to include governmental agencies not now involved which have major missions

involving lifelines and non-governmental activities. Present governmental members

on each side to propose details of such expansions In their respective countries.

519
2) Coordination effort of U.S. -Japan joint research program on the safety

of lifeline system and its components at every level of cooperation should be

initiated through the co-chairman of the Task Committee.

3) Planning of cooperative large-scale testing of lifeline systems should

immediately be initiated, with details to be arranged by correspondence.

A) In preparation for items 2) and 3) mentioned above, members of the Task

Committee on both sides should be expanded to the extent of covering each field

of lifelines concurred by both co-chairmen as subjects of cooperative research

program.

520
(G) Wind

UJNR

Report of Task Committee on

Characteristics and Structural Response

PANEL ON WIND AND SEISMIC EFFECTS

Date; Hay 19, 1980

Place : National Bureau of Standards

Washington, D.C. U.S.A.

Attendees : U.S. Side - Celso S. Barrientos (Chairman)

Richard D. Marshall

Japan Side - Toshio Iwasaki (Chairman, Temporary)

Hajime Tsuchida

Yasuyuki Koga

To expand the activities of the Task Committee, the name of the Task

Committee was changed to "Wind Characteristics and Structural Response,"

and two members will be added on the U.S. side.

(1) Activities and Principal Accomplishments to Date

The Task Committee notes that:

1) Strong winds frequently cause loss of life and extensive property

damage

,

2) The exchange of high wind data and information on wind effects

can be useful for reducing loss of life and property damage,

3) Insufficient knowledge exists on the effects of strong winds upon

structures and methods for modeling the:je effects, and

4) The needs for wipd data for various application:; and new projects

are increasing.

( 2 ) Future Activiti es

In view of the above Items, the Task Committee hereby resolves to

carry out the following programs:

521
Exchange observations and records of high wind data (including

remote-sensing reports, i.e., aircraft, satellite) available from

the respective national meteorological services and special

observation sites.

Encourage the exchange cf actual wind observations (described in 1)

in appropriate format. When necessary, the Task Committee Chairman

will establish procedures to facilitate the exchange of such data.

Exchange available specialized high wind data sets and wind pressure

sets. The information may include surface meteorological data sets and

data sets at other levels. Documentation of instrument characteristics,

exposure, and elevation above ground may be included.

Enccurage the establishment of standard methods for the simulation

of boundary layers in wind tunnels, the promotion of data exchange

of boundary layer wind observations, and the exchange of information

on methods cf measurement and determination of structural response.

Encourage thf. interaction between meteorologists and engineers to

identify the types of wind c'ata required for future use; in establishing

extreme wind distribution;;; in determining wind loadings on structures ;

ir understanding the urban wind climate; and in considering structural

design issues in the wind generation of energy.

522
Report of the Task Committee On

(H) Soil Behavior and Stability During Earthquakes

UJNR Panel on Wind and Seismic Effects

Date : May 21, 1980

Place : Bureau of Standards, Gaithersburg , Maryland, U.S.A.

Attendees: Japan side - Toshio Iwasaki (Chairman)

Hajime Tsuchida

Yasuyuki Koga

U.S. side - H. S. Lew (Temporary Chairman)

Felix Y. Yokel

Riley M. Chung

(1) Activities and Principal Accomplishments to Date :

1) Pursuant to the resolution of the task committee of the 11th

Joint Meeting, the co-chairman of the task committee prepared

mailing lists and exchanged them for the distribution of documents .

2) Exchanges of relevant documents, including reports on the

recent earthquakes and guidelines for assessing soil liquefaction were made.

(2) Future Programs

1) When requested, the task committee will assist in arrangements

for visits to earth embankments and soil foundations which have

been subjected to significant ground motions.

2) Discussions and correspondence are being continued concerning

soil liquefaction potential for submerged foundations, foundation settlements, and stability of soil structures due to earthquake motions.

3) Efforts for comparing results of standard penetration tests

conducted in the United States with the results from Japan

will be continued.

A) Guidelines for assessing soil liquefaction in the United States

and Japan are currently being reviewed and final drafts are ex- pected to be published in Che near future. These reports will be

transmitted to counterpart sides when completed.

523
Report of Task Committee on

(I) Storm Surge and Tsunami

UJNR PANEL ON WIND AND SEISMIC EFFECTS

Date : May 21, 1980

Place : National Bureau of Standards

Washington, D.C. U.S.A.

Attendees : U.S. side - Celso S. Barrientos (Chairman)

Japan side - Haj ime Tsuchida (Chairman, Temporary)

This is the newest Task Committee of this Panel. The Task Committee

was formally formed during the 11th Panel Meeting in Tsukuba, Japan in

September 1979. Members of the Task Committee were appointed as follows:

U.S. side - Celso S. Barrientos (Chairman)

Michael P. Gaus

Japan side - Yoshimi Goda (Chairman)

Hiroshi Hasimoto

Akira Katayama

It is desirable to add memberships in the Committee: U.S. side -

tsunami research and tsunami forecasting.

Storm surges and tsunamis caused considerable damage to properties

and loss of lives in Japan and the United States and in other countries where

these geophysical phenomena occur. Storm surge and tsunami are clearly

wind and seismic effects and therefore a very logical part of this Panel.

The Task Committee will facilitate exchange of information related to

storm surge and tsunami, such as, publications, damages, and measures to

reduce or prevent damages.

5Z4
The Chairman of the Task Committee will consider the following

peration programs:

Exchange information on storm surge models used for forecasting,

planning, design, and disaster prevention purposes. The information

include type and formulation of models, verification data, and

empirical constants. Contacts between scientists working on storm

surge models will also be part of the information exchange.

Encourage development of meteorological models that are aimed to be

utilized with the storm surge models. These are atmospheric models

that are designed as input initial conditions to the surge models.

Since storm surge generation is primarily dependent on the atmospheric

forcings, storm" surge models are only as good as the atmospheric

models

.

525
NBS-114A ipev. 2-aci U.S. DEPT. OF COMM. 1. PUBLICATION OR 2. Performing Organ. Report No. 3. Publ i cation D ate BIBLIOGRAPHIC DATA

REPORT NO. SHEET (See instructions) NBS SP 655 January 1984

4. TITLE AND SUBTITLE

Wind and Seismic Effects: Proceedings of the 12th Joint UJNR Conference

5. AUTHOR(S)

E. V. Leyendecker and R. M. Chung, editors

6. PERFORMING ORGANIZATION (If joint or other than NBS. see instructions)

NATIONAL BUREAU OF STANDARDS

DEPARTMENT OF COMMERCE

WASHINGTON, D.C. 20234

7. Contract/Grant No.

8. Type of Report & Period Covered

Final

9. SPONSORING ORGANIZATION NAME AND COMPLETE ADDRESS (Street. City. State, ZIP)

Same as item 6.

10. SUPPLEMENTARY NOTES

Library of Congress Catalog Card Number: 83-600593

] Document describes a computer program; SF-185, FlPS Software Summary, is attached.

11. ABSTRACT (A 200-word or less factual summary of most si gnificant information . If document includes a significant

bi bl 10 graphy or literature survey, mention it here)

The Twelfth Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects was held

in Gaithersburg , Maryland on May 19-23, 1980. The proceedings of the Joint Meeting

include the program, the formal resolutions, the Task Committee Reports, and the technical papers. The subjects covered in the papers include: (1) the characterization of

seismic ground motion, (2) the characterization of natural wind and extreme wind

records, (3) structural response to earthquake loading, (4) storm surge and tsunamis,

(5) recent developments in seismic design criteria, (6) technical cooperation with

developing countries, (7) earthquake hazard mitigation, and (8) structural response to

wind loading.

12. KEY WORDS (Six to twelve entries; alphabetical order; capitalize only proper names; and separate key words by semi colon s) accelerograph ; bridges; codes; design criteria; disaster; earthquakes; geotechnical

engineering; ground failures; seismic design; seismicity; standards; structural

engineering; tsunamis; and winds.

13. AVAILABILITY

[XJ Unl imited For Official Distribution. Do Not Release to NTIS

Order From Superintendent of Documents, U.S. Government Printing Office, Washington, D.C.

~ 20402.

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14. NO. OF

PRINTED PAGES

543

15. Price

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