NAT L INST. Of ST4N0 S TECH R.I.C. NIST

PUBLICATIONS

A111D3

1^ 1 1 C^rr ^"'^^^ states Department of Commerce

I 1 National Institute of Standards and Technology

NIST Special Publication 778 (ICCSSC TRll)

Performance of Structures During the

Loma Prieta Earthquake of October 17y 1989

H. S. Lew, Editor
NATIONAL INSTITUTE OF STANDARDS &

TECHNOLOGY

Research Information Center

Gaithersburg, MD 20899
NIST Special Publication 778

Performance of Structures During the

Loma Prieta Earthquake of October 1 7y 1989

H. S. Lew, Editor

Center for Building Technology

National Engineering Laboratory

National Institute of Standards and Technology

Gaithersburg, MD 20899

Sponsored by:

U.S. Geological Survey

Department of the Interior

Reston, VA 22092

January 1990

U.S. Department of Commerce

Robert A. Mosbacher, Secretary

National Institute of Standards and Technology

Raymond G. Kammer, Acting Director
National Institute of Standards U.S. Government Printing Office

and Technology Washington: 1990

Special Publication 778

Natl. Inst. Stand. Technol.

Spec. Publ. 778

201 pages (Jan. 1990)

CODEN: NSPUE2

For sale by the Superintendent

of Documents

U.S. Government Printing Office

Washington, DC 20402
ABSTRACT

Immediately following the magnitude 7.1 Loma Prieta earthquake of October 17, 1989, a team representing

the Interagency Committee on Seismic Safety in Construction surveyed the damage to buildings, utilities

and transportation structures. This report is based primarily on the data gathered during the site survey

and the results of preliminary analyses of structural failures. Most structures designed in accordance with

modern codes and standards performed well without serious structural damage. However, there were many

concrete and masonry buildings and highway structures in the San Francisco Bay area which were not

designed according to modern seismic design codes and which did not perform well. The majority of

damaged structures had not been strengthened to increase earthquake resistance. Except for two deaths

from landslides and one from fire, the remainder of the 62 deaths from the earthquake was due to partial

or total collapse of older structures. This investigation provided the basis for recommendations to improve

design and construction practices for buildings and lifeline structures and to mitigate damage to existing

structures in future earthquakes.

Key Words: Bridges; buildings; codes; damage; earthquakes; fire safety; geology; highways; housing;

lifelines; seismic; specifications; structural engineering; viaducts.

iii

PREFACE

The Loma Prieta earthquake struck the San Francisco Bay region at 5:04 p.m. (PDT) on October 17, 1989.

Within a few hours, a decision was made to dispatch an interagency team representing the Interagency

Committee on Seismic Safety in Construction (ICSSC). ICSSC is comprised of representatives of 25

Federal agencies with responsibilities for seismic safety of Federal or Federally assisted facilities. On

October 18, 1989, the U.S. Senate Committee on Commerce, Science and Transportation and the U.S.

House of Representatives Committee on Science, Space and Technology, Subcommittee on Science,

Research and Technology requested the National Institute of Standards and Technology (NIST) to investigate earthquake damage including the elevated section of Interstate 880 (the Nimitz Freeway) and

other bridge structures. Since NIST provides the secretariat for ICSSC, the team responded to the

Congressional request. The team arrived in the San Francisco area on October 18, 1989, and carried out

its investigation during the period of October 19 - 26, 1989. The team included civil, fire safety,

geotechnical, and structural engineers. The team members were:

National Institute of Standards and Technology

Nicholas J. Carino

H. S. Lew, Team Leader

Harold E. Nelson

William C. Stone

Richard N. Wright

Felix Y. Yokel

Federal Highway Administration

James D. Cooper

Department of Housing and Urban Development

Lincoln M. H. Chang

U.S. Geological Survey

E. V. Leyendecker

The primary purposes of this report are to document perishable data and to identify opportunities to gain

knowledge to improve practices for the seismic safety of new and existing buildings and lifelines.

The author or authors of each of the chapters are:

Chapter 1. H. S. Lew

Chapter 2. E. V. Leyendecker, Paul C. Thenhaus, Kenneth W.

Campbell, Margaret G. Hopper, Stanley L. Hanson,

S. T. Algermissen and David M. Perkins (USGS

contributors)

Chapter 3. Felix Y. Yokel

Chapter 4. Nicholas J. Carino

Chapter 5. William C. Stone, James D. Cooper and Nicholas J. Carino

Chapter 6. Harold E. Nelson

Chapter 7. Richard N. Wright and H. S. Lew

V
ACKNOWLEDGMENTS

The team acknowledges the logistical support provided by the Federal Emergency Management Agency's

Region IX office. Ms. Laurie Friedman, who was assigned to the team by FEMA, made arrangements for

the team to visit several sites in San Francisco and the epicentral region.

James H. Gates of the California Department of Transportation arranged the team's visit to the 1-880

and other damage sites and made available the drawings of the bridge and highway structures. Frank

Lew of the city of San Francisco guided the team through the Marina District in inclement weather and

provided the building damage statistics. Ken Lewis, Senior Civil Engineer, city of Watsonville, provided

subsurface information obtained from the Watsonville area.

Carl Stover and Glen Reagor, USGS, assisted in the interpretation of Modified Mercalli Intensity data

collected from a mail survey for inclusion in Chapter 2. Phil Powers assisted with computer analysis of

data and preparation of figures.

Many of the illustrations in this report were prepared by Keith F. Mackley of NIST.

The preparation of this report was partially supported by the U.S. Geological Survey, Department of the

Interior.

vi
CONTENTS

ABSTRACT iii

PREFACE V

ACKNOWLEDGMENTS vi

1. INTRODUCTION 1-1

2. SEISMICITY AND STRONG MOTION DATA 2-1

2.1 Introduction 2-1

2.2 Earthquake Mechanism 2-1

2.3 Strong Ground Motion 2-2

2.4 The Modified Mercalli Scale and MM Assessments 2-3

2.5 Summary 2-6

2.6 References . i 2-7

3. SURFICIAL GEOLOGY AND FOUNDATION FAILURES 3-1

3.1 Introduction 3-1

3.2 Ground Failures 3-1

3.3 Relation of Subsurface Conditions to Earthquake Damage 3-2

3.3.1 San Francisco 3-2

3.3.2 Oakland 3-5

3.3.3 Watsonville 3-6

3.3.4 Santa Cruz 3-7

3.3.5 Los Gatos 3-8

3.4 Summary 3-8

3.5 References 3-9

4. PERFORMANCE OF BUILDINGS 4-1

4.1 Introduction 4-1

4.2 San Francisco 4-1

4.2.1 Marina District 4-2

4.2.2 Financial, South of Market, and Civic Center Districts 4-4

4.3 City of Oakland 4-33

4.4 Epicentral Region 4-41

4.5 Summary 4-43

5. PERFORMANCE OF BRIDGE AND HIGHWAY STRUCTURES 5-1

5.1 Introduction 5-1

5.2 Historical Development of Seismic Criteria 5-1

5.3 Current Seismic Design Criteria 5-2

5.4 Approach for Improved Seismic Response 5-4

5.5 Introduction to Viaducts and Bridge Performance 5-5

5.6 The 1-880 Collapse 5-7

5.6.1 Description of Structure 5-7

5.6.2 Description of Collapse by Motorists 5-11

5.6.3 Summary of Aerial Photos 5-12

5.6.4 Ground Details 5-13

5.6.5 Observations on Contributing Causes of Collapse 5-16

vii
5.6.6 Summary 5-20

5.7 Bay Bridge 5-58

5.8 Struve Slough Bridges 5-69

5.9 Interstate 280 5-71

5.10 Embarcadero Freeway (1-480) 5-83

5.11 U.S. Highway 101 5-89

5.12 References 5-92

6. PERFORMANCE OF FIRE PROTECTION SYSTEMS 6-1

6.1 Introduction 6-1

6.2 General Observations 6-1

6.3 Meteorological Effects 6-2

6.4 Private Fire Protection Systems 6-3

6.5 Summary 6-3

7. CONCLUSIONS AND RECOMMENDATIONS 7-1

7.1 Conclusions 7-1

7.1.1 Performance of All Structures 7-1

7.1.2 Buildings Performance 7-3

7.1.3 Lifelines Performance 7-3

7.2 Recommendations 7-4

7.2.1 Performance of All Structures 7-4

7.2.2 Buildings Performance 7-5

7.2.3 Lifelines Performance 7-6

7.3 References 7-7

viii
1. INTRODUCTION

by H. S. Lew

At 5:04 p.m., Pacific Daylight Time, on October 17, 1989, an earthquake with a surface-wave magnitude

of 7.1 occurred with its epicenter located about 10 miles (15 km) northeast of Santa Cruz and 60 miles

(95 km) south-southeast of San Francisco, California. According to the U.S. Geological Survey, the

earthquake ruptured a segment of the San Andreas fault below the Santa Cruz Mountains. The hypocenter

was about 11 miles (18 km) beneath the Earth's surface, and the rupture propagated about 25 miles (40

km) both northwest and southeast within a 10-second period. The earthquake was felt over an area of

400,000 square miles (1,000,000 sq km), from Los Angeles to the south, Oregon to the north, and western

Nevada to the east. This earthquake, named the Loma Prieta earthquake, was the largest on the San

Andreas fault since the great San Francisco earthquake of 1906 (M = 8.3) when a 275-mile (440-km)

stretch of the fault ruptured.

Strong shaking lasted only about 10 to 15 seconds. Even so the destructive effects of the Loma Prieta

earthquake were extensive. Wood-frame dwellings and unreinforced masonry buildings in communities near

the epicenter and in isolated locations in San Francisco and Oakland sustained substantial damage. In

Santa Cruz, Watsonville, and Los Gatos, there was extensive damage to old buildings in the downtown

areas where small businesses are located. Elevated highway structures sustained severe damage in the Bay

area, including the collapse of a section of Interstate 880 in Oakland which claimed 42 lives. The collapse

of a 50-foot (15-m) link span of the San Francisco-Oakland Bay Bridge caused the bridge to remain closed

for about a month. The Bay Area Rapid Transit System was in service immediately after the earthquake.

However, damage to arterial highways and the bridge caused severe impacts on commuters and regional

commerce.

As of December 28, 1989, the following statistics have been compiled by the California Governor's Office

of Emergency Services:

— 62 confirmed deaths; 3,757 injuries. — Over $6 billion property damage. — Over 12,000 people displaced from their homes immediately following the

earthquake.

This report presents an overview of the nature of the Loma Prieta earthquake and its effects and

documents the damage. The geotechnical engineering aspects of the affected region are discussed in detail

1-1
as they affect in large measure the distribution of the observed damage. Because of the significance of

the damage to highway structures, the evolution of the design criteria for these structures is described.

The collapse of the elevated portion of Interstate 880 is discussed in detail and preliminary analyses to

determine the causes of the failure are presented. A detailed description of the nature of the damage

sustained by buildings located in the Bay area and in the epicentral region is provided. The extent of fire

damage and the performance of fire protection systems are presented. The lessons learned, or re-learned,

from this earthquake are summarized. The report concludes with recommendations for actions to reduce

risks from future earthquakes and for research to exploit opportunities provided by the Loma Prieta

earthquake for learning and improving practices.

1-2
2. SEISMICITY AND STRONG MOTION DATA

by Edgar V. Leyendecker, Paul C. Thenhaus,

Kenneth W. Campbell, Margaret G. Hopper, Stanley L. Hanson,

S. T. Algermissen, and David M. Perkins

2.1 Introduction

The October 17, 1989 Loma Prieta earthquake (M, 7.1) was the largest earthquake to strike the San

Francisco Bay area since the great San Francisco earthquake of 1906. Numerous investigations were

initiated as a consequence of this earthquake. One of these investigations involved collection of Modified

Mercalli (MM) intensity data and collection of damage statistics for loss estimation research (see

Leyendecker and others, 1988, for a description of field procedures). These data and data collected by an

investigation team from the Interagency Committee in Seismic Safety in Construction formed part of a data

base that led to the preliminary interpretation of the distribution of Modified Mercalli intensities presented

here. Strong motion records collected by the U.S. Geological Survey (USGS) and the State of California's

Division of Mines and Geology, provided basic ground-motion data that are discussed in conjunction with

the intensity data.

2.2 Earthquake Mechanism

The epicenter was located 10 miles (15 km) northeast of Santa Cruz and was 60 miles (95 km) southsoutheast of San Francisco at 37.036° N. latitude, 121.883° W. longitude (Ward and Page, 1989).

Earthquake rupture initiated at 5:04:15 p.m.. Pacific Daylight Time, at a depth of 11.4 miles (18.4 km) on

the Southern Santa Cruz Mountains segment of the San Andreas fault (fig. 2.2.1(a)). Coseismic rupture

propagated up-dip to a depth of 3.7 miles (5.9 km) and bilaterally along the fault northwest to just south

of Los Gatos, and southeast to a point just south of Watsonville for a rupture length of 25 miles (40 km).

From the south, the San Andreas fault bends to a more northwesterly course through this segment and

results in a high degree of compression across the right-slip fault system. The fault plane dipped

approximately 70° to the southwest and had a strong west-over-east thrust component (fig. 2.2.1(b)).

Geodetic surveys following the earthquake indicate that slip at the hypocenter had a right-lateral horizontal

component of 6.2 feet (1.9 m) and a vertical component of 4.3 feet (1.3 m) that resolve to a total obliqueslip displacement of 7.5 feet (2.3 m) (Plafker and Galloway, 1989). Deformation at the surface consisted

of 14 inches (0.35 m) of uplift southwest of the fault's surface trace and echelon tensional cracks in the

uplifted southwest block (fig. 2.2.1(c)). Surface fault displacement has not yet been identified

2-1
unequivocally, but the lack of surface fault displacement would be consistent with the top of the rupture

zone being at a depth of about 4 miles (6 km).

The southern Santa Cruz Mountains segment of the San Andreas fault was identified in 1988 as having

a 30 percent probability of rupturing in a magnitude 6.5 earthquake within the next 30 years (U.S.

Geological Survey, 1988). Figure 2.2.2 is a cross section of seismicity along the San Andreas fault as

recorded during the past 20 years for an area extending from just south of Parkfield to north of San

Francisco. Much of the fault zone between Parkfield and San Juan Batista is characterized by a high rate

of low-magnitude seismicity associated with the well documented creep mechanism of fault slip in this

segment. North and south of this area, however, seismicity is distributed both horizontally and vertically

in fairly discrete patches along the fault plane. Areas deficient in earthquakes are areas of assumed strain

accumulation and, therefore, are considered the most likely areas of major earthquakes in the near future.

Three such seismic gaps are labeled in figure 2.2.2(a). Figure 2.2.2(b) illustrates how the Loma Prieta

earthquake and its aftershocks filled a previously recognized seismic gap.

2.3 Strong Ground Motion

The earthquake triggered over 130 strong-motion instruments operated by the USGS (Maley et al., 1989)

and the California Division of Mines and Geology (Shakal et al., 1989). Station locations are shown in

figure 2.3.1. The amplitude of the larger of the two peak horizontal components of acceleration is shown

adjacent to each station (units are fraction of gravity). The stations in the vicinity of San Francisco and

Oakland are shown more clearly in figure 2.3.2. The larger of the two peak horizontal acceleration

components at each station are shown graphically in figure 2.3.3 with the acceleration amplitude in

proportion to the diameter of the circle.

The accelerograms recorded at 10 strong-motion stations, along with their peak accelerations, epicentral

distances, and geologic site conditions, are shown in figure 2.3.4. The location of these 10 stations are

identified by an alphabetic designation (A-J) in figures 2.3.1, 2.3.2, and 2.3.3.

The duration of strong shaking for two rock sites located over the inferred rupture zone of the earthquake

(e.g., Lexington Dam, station D, and Corralitos, station A) is approximately 5 to 7 seconds. This duration

is consistent with bilateral rupture over a total distance of about 25 miles (40 km), and with interpretations

of the geodetic data. The estimated 25 mile (40 km) rupture length almost exactly coincides with the

geographical limits of the MM intensity VIII contour inferred for the epicentral region (fig. 2.3.5).

2-2
Within the epicentral source region, peak accelerations were observed to be relatively independent of

surface geology. However, outside of this region (especially beyond distances of 50 km or so), surface

geology appears to strongly influence the amplitude of ground motion: sites located on crystalline rock

and rocks of the Franciscan Complex have the lowest accelerations, sites located on soft rock and alluvium

have intermediate accelerations, and sites located on artificial fill and Bay Mud have the highest

accelerations. The differences in horizontal acceleration at hard-rock and Bay-Mud sites were best

demonstrated in the San Francisco and Oakland areas, where differences of 100% to 260% were observed.

Within the epicentral source region of the earthquake, many sites recorded nearly equal amplitudes for

the vertical and horizontal components of peak acceleration. This is consistent with limited observations

from past earthquakes, having a strike-slip mechanism.

Peak acceleration was observed to be strongly dependent on geographical area and azimuth (see fig. 2.3.3),

especially outside of the near-source region (about 30 miles (48 km) as defined by Campbell, 1981, 1989).

For example, sites underlain by a given surface geology located to the northwest of the epicenter in the

direction of San Francisco and Oakland had consistently higher accelerations than sites located to the north

of the epicenter in the direction of San Jose, Hayward, and Livermore Valley. This observation is

consistent with the overall distribution of MM intensities shown in figure 2.3.5. The combined effects of

source directivity and radiation pattern may be partly responsible for this pattern, although some bias may

be introduced to the overall damage pattern by the northwest trending corridor of cultural development

along major highways. Also, the instrumentation array is relatively sparse in the Diablo Range to the

northeast and in the western Santa Cruz Mountains southwest of the epicenter.

2.4 The Modified Mercalli Scale and MM Assessments

The Modified Mercalli intensity (MM) scale of 1931 (Wood and Neuman, 1931) groups observations on

earthquake effects into similar levels of shaking and then ranks the shaking levels into ascending order I

through XII (see table 2.1). Ground failure phenomena, such as landslides, sand blows, rock falls and

liquefaction are generally relegated to MM level IX and higher. However, research and observations since

the scale was developed in 1931 show that such effects can occur at much lower intensity levels depending

on a number of physical properties of the surficial materials, such as moisture content, permeability, and

texture, and the slope angles of hillsides and bluffs. Accordingly, where possible, the MM intensities have

been assigned on the basis of shaking damage to buildings and structures and not on the basis of ground

failure effects. This exercise becomes difficult, however, where structural damage from ground failure and

shaking are inextricably mixed, as was the case for certain structures located in the communities of

2-3
Redwood Estates, Santa Cruz and the Marina District of San Francisco. In these cases, assessments of

nearby structures and other items such as water towers and telephone poles, apparently not influenced by

local ground failure effects, were used to corroborate overall assessments. However, damage resulting from

the possible amplification of ground motion by surficial geologic materials and local topography is

legitimately classed as shaking damage in these assessments, since amplification is believed to be unrelated

to ground failure.

Areas assigned MM VI are characterized by broken windows and minor cracking to unreinforced masonry

chimneys and plaster walls. For convenience, MM VII can be characterized as the threshold of architectural

damage to buildings; negligible damage to well-designed and built ordinary buildings but considerable

damage to poorly designed and built buildings. Falling of weak chimneys at the roof-line is characteristic

of MM VII damage. MM VIII is characterized generally as the threshold of structural damage to

buildings. Damage is considerable to ordinary substantial buildings. Some wood-framed houses are racked

and move or fall off of foundations. Such damage at this level is usually because 1) the houses are

unattached to the foundation, 2) the pile-foundation or wooden sills resting on a masonry or concrete

foundation are badly decayed, or 3) there is insufficient lateral bracing of cripple studs which support the

floor joists of the first story. Unreinforced masonry walls fall. MM IX is marked by considerable damage

to structures designed to withstand earthquakes.

Figure 2.3.5 shows a preliminary map of the regional distribution of Modified Mercalli intensity resulting

from the Loma Prieta earthquake. More detailed assessments in the San Francisco and Oakland areas are

shown in the map in figure 2.4.1. Considerable work remains, however, to finalize the details of this

initial assessment. Except for the MM VI assessments in Brentwood, Banta, Manteca, Vernalis, and some

areas along state Highway 1, the evaluations are based on primary observations and data collected by the

field parties. A secondary source of information for the above locations was early responses to a mail

survey of postmasters, police and fire departments conducted by the U.S. Geological Survey's National

Earthquake Information Center.

While the epicentral intensity of the Loma Prieta earthquake was assigned an MM VIII based on

substantial damage to wood-framed dwellings and unreinforced masonry buildings in communities near the

epicenter, the highest intensity levels (MM IX) were assigned to isolated sites in San Francisco and

Oakland. The collapse of the elevated portion of 1-880 in Oakland and the considerable damage to the

Embarcadero Freeway in San Francisco warrant MM IX. Both of the reinforced concrete freeway

structures were built under seismic design requirements of the then-existing codes. The Marina District

in northern San Francisco also was assigned MM IX. Ground failure and shaking both played a role in

2-4
some apartment collapses in the district. However, other collapses occurred in areas of no apparent

ground failure. The collapse and widespread structural damage to these substantial buildings is reason

for a tentative assignment of MM IX to the Marina District.

In all of the above areas, amplification of ground motions by surficial geologic materials may have played

a significant role in causing the observed damage. Notably, intensity levels in eastern and northern San

Francisco (known areas of thick sediments and Bay Mud; Joyner, 1982) exhibit intensity levels one to three

units higher than the central-city areas. An exception is the extreme western margin of San Francisco

where thick sediments have been documented but intensity levels are no higher than in the central area

(fig. 2.4.1). These large differences in MM intensity are also consistent with the observed 100% to 200%

difference in peak horizontal accelerations recorded on Bay Mud and Franciscan rock sites in the San

Francisco and Oakland areas.

In the MM VIII epicentral area, wood-framed houses on laterally unbraced cripple studs performed poorly,

as was the case in the May 2, 1983 Coalinga, California earthquake (Hopper et al., 1983). In Los Gatos,

stately refurbished Victorian houses were thrown off their foundations. Decayed and rotted sills at the

foundation connection of these older homes indicate a need for structural rehabilitation as well as cosmetic

remodelling. Retrofitting decayed sills at the foundation level and laterally bracing cripple studs below the

first-floor level would help to mitigate similarly extreme residential damage expected in future earthquakes.

It is noted that many residences of post- 1950 construction, when improved anchoring to the foundations

began to be required, sustained far less damage than residences of earlier periods of construction.

The MM intensities along state Highway 1 on the Pacific Coast were assigned values of 6 and 7. Thus,

the closure of the isoseismal between MM VI and MM VII between Monterey and Half Moon Bay seems

likely to be located at or near the coastline (fig. 2.3.5).

The maps in figures 2.3.5 and 2.4.1 were originally developed by the authors and published by the USGS

in a Circular on the Loma Prieta earthquake (Plafker and Galloway, 1989). Additional data obtained since

the original publication indicate the need for some modifications. However, rather than modify the maps

at this time on the basis of incomplete data, some of the additional data are shown highlighted on the

maps to indicate trends.

As an example of the wealth of data available, figure 2.4.2 shows some of the safety inspection results

available for San Francisco (Lew, 1989). These represent 378 inspected buildings using post-earthquake

safety evaluation procedures developed by the Applied Technology Council (ATC, 1989). These are only

2-5
for buildings posted as unsafe to enter (posted "Red" by the ATC procedures). The preliminary contour

is shown for comparison. Additional data are available for San Francisco and other cities, such as

Oakland, and will be used to modify the isoseismals in the future.

2.5 Summary ^

In the epicentral region of the earthquake, many sites recorded nearly equal amplitudes for the vertical

and horizontal components of peak acceleration. This is consistent with observations of past strike-slip

earthquakes. The peak horizontal acceleration of 0.64g was measured in the epicentral region at

Corralitos. The peak vertical acceleration at this location was 0.47g. Duration of strong ground motion

in the epicentral area was on the order of 5 to 7 seconds and 10 to 15 seconds in San Francisco and

Oakland. The inferred rupture length of 25 miles (40 km) almost coincides with the geographical limits

of the MM intensity VIII contour for the epicentral region.

/

Peak accelerations were observed to be strongly dependent on geographical area and azimuth, especially

outside of the near-source region. For example, sites of a given surface geology located to the northwest

of the epicenter in the direction of San Francisco and Oakland had consistently higher accelerations than

sites located to the north of the epicenter in the direction of San Jose, Hayward, and Livermore Valley.

This observation is also consistent with the overall distribution of MM intensities.

The epicentral intensity of the Loma Prieta earthquake was assigned an MM VIII based on substantial

damage to wood-framed dwellings and unreinforced masonry buildings in communities near the epicenter.

The highest intensity levels (MM IX) were assigned to isolated sites in San Francisco and Oakland some

55 to 60 miles (90 to 100 km) from the epicenter. Amplification of ground motions by surficial geologic

materials may have played a significant role in causing these high intensities at such large distances.

Notably, intensity levels in eastern and northern San Francisco, known areas of thick sediments and Bay

Mud, exhibited intensity levels one to three units higher than the central San Francisco area. The large

differences in MM intensity are consistent with the observed differences in peak accelerations. In the San

Francisco-Oakland area, differences of 100% to 260% were observed in peak horizontal accelerations

recorded on Bay Mud and Franciscan rock sites.

In the MM VIII epicentral area, wood-framed houses on laterally unbraced cripple studs performed poorly,

as was the case in past earthquakes. Many residences of post- 1950 construction, when improved anchoring

to the foundations began to be required, sustained less damage than residences of earlier periods of

construction.

2-6
2.6 References

Applied Technology Council. 1989a. Procedures for postearthquake safety evaluation of buildings. Applied

Technology Council, ATC-20. 152 p.

Applied Technology Council. 1989b. Field Manual: Postearthquake safety evaluation of buildings. Applied

Technology Council, ATC-20-1. 114 p.

Campbell, K. W. 1981. Near-source attenuation of peak horizontal acceleration. Bull, of the Seis. Soc. of

Am. 68:828-843.

Campbell, K. W. 1989. Empirical prediction of near-source ground motion for the Diablo Canyon power

plant site, San Luis Obispo County, California. U.S. Geological Survey Open-File Report 89-484, 115 p.

Hopper, M. G., Thenhaus, P. C, Barnhard, L. M., and Algermissen, S. T. 1983. Damage survey in

Coalinga, California for the earthquake of May 2, 1983. The 1983 Coalinga, California earthquake.

California Department of Conservation, Division of Mines and Geology. Special Publication 66, p. 5-8.

Joyner, W. B. 1982. Map showing the 200-foot thickness contour of surficial deposits and the landward

limit of bay mud deposits of San Francisco, California: U.S. Geological Survey Miscellaneous Field Studies

Map MF-1376, 1 plate, scale 1:24,000.

Lew, Frank. 1989. Department of Public Works, City and County of San Francisco, Personal

communication with reports on earthquake-related building inspections.

Leyendecker, E. V., Highland, L. M., Hopper, M. G., and Arnold, E. P., Thenhaus, P. C, and Powers, P. 1988. The Whittier Narrows, California earthquake of October 1, 1987~Early results of isoseismal studies

and damage surveys. Earthquake Spectra, v. 4, p. 1-10.

Maley, R., Acosta, A., Ellis, F., Etheredge, E., Foote, L., Johnson, D., Porcella, R., Salsman, M., and

Switzer, J. October 1989. Strong-motion records from the Northern California (Loma Prieta) earthquake

of October 17, 1989. U.S. Geological Survey Open-File Report 89-568. 85 p.

Plafker, G. and Galloway, J. 1989. Lessons learned from the Loma Prieta, California earthquake of October

17, 1989. U.S. Geological Survey Circular 1045. 48 p.

Shakal, A., Huang, M., Reichle, M., Ventura, C, Cao, T., Sherburne, R., Savage, M., Darragh, R., and

Peterson, C. November 1989. CSMIP strong-motion records from the Santa Cruz Mountains (Loma Prieta),

California earthquake of 17 October 1989. California Division of Mines and Geology, California Strong

Motion Instrumentation Program Report No. OSMS 89-06. 196 p.

Stover, Carl. 1989. U.S. Geological Survey. Personal communication on evaluation of responses to an

incomplete mail survey of MMI.

U.S. Geological Survey. 1988. Probabilities of large earthquakes occurring in California on the San Andreas

fault. U.S. Geological Survey Open-File report 88-398. 62 p.

Ward, P. and Page, R. 1989. The Loma Prieta earthquake of October 17, 1989. U.S. Geological Survey

pamphlet. 16 p.

Wood, H. O., and Neuman, F. 1931. Modified Mercalli intensity scale of 1931. Seismological Society of

America Bulletin, v. 21. p. 277-283.

2-7
Table 2.1 Modified Mercalli intensity scale of 1931 (Wood and Neuman, 1931).

I Not felt, or, except rarely under especially favorable circumstances.

Under certain conditions, at and outside the boundary of the area in which a great shock is felt: sometimes birds, animals, reported uneasy

or disturbed; sometimes dizziness or nausea experienced; sometimes

trees, structures, liquids, bodies of water, may sway-doors may swing,

veiy slowly.

II Felt indoors by few, especially on upper floors, or by sensitive or

nervous persons. Also, as in grade I but often more noticeably:

sometimes hanging objects may swing, especially when delicately

suspended; sometimes trees, structures, liquids, bodies of water, may

sway, doors may swing very slowly; sometimes birds, animals, reported uneasy or disturbed; sometimes dizziness or nausea experienced.

III Felt indoors by several, motion usually rapid vibration. Sometimes not recognized to be an earthquake at first. Duration

estimated in some cases. Vibration like that due to passing of light, or lightly loaded trucks, or heavy trucks some distance away. Hanging objects may swing slightly. Movements may be appreciable on upper levels of tall structures. Rocked standing motor cars

slightly.

rv Felt indoors by many, outdoors by few. Awakened few, especially

light sleepers. Frightened no one, unless apprehensive from previous

experience. Vibration like that due to passing of heavy, or heavily

loaded trucks. Sensation like heavy body striking building, or falling of heavy objects inside. Rattling of dishes, windows, doors; glassware

and crockeiy clink and clash. Creaking of walls, frame, especially in the upper range of this grade. Hanging objects swung, in numerous

instances. Disturbed liquids in open vessels slightly. Rocked standing

motor cars noticeably.

V Felt indoors by practically all, outdoors by many or most; outdoors

direction estimated. Awakened many, or most. Frightened few slight

excitement, a few ran outdoors. Building trembled throughout.

Broke dishes, glassware, to some extent. Cracked windows in some

cases, but not generally. Overturned vases, small or unstable objects

in many instance, with occasional fall. Hanging objects, doors, swing

generally or considerably. Knocked pictures against walls, or swung

them out of place. Opened, or closed, doors, shutters, abruptly.

Pendulum clocks stopped, started, or ran fast, or slow. Moved small

objects, furnishings, the latter to slight extent. Spilled liquids in small amounts from well-filled open containers. Trees, bushes, shaken

slightly.

VI Felt by all, indoors and outdoors. Frightened many, excitement

general, some alarm, many ran outdoors. Awakened all. Persons made to move unsteadily. Trees, bushes, shaken slightly to moderately. Liquid set in strong motion. Small bells rang-church,

chapel, school, etc. Damage slight in poorly built buildings. Fall of

plaster in small amount. Cracked plaster somewhat, especially fine cracks, chimneys in some instances. Broke dishes, glassware, in considerable quantity, also some windows. Fall of knickknacks, books,

pictures. Overturned furniture in many instances. Moved furnishings of moderately heavy kind.

VII Frightened all, general alarm, all ran outdoors. Some, or many,

found it difficult to stand. Noticed by persons driving motor cars. Trees and bushes shaken moderately to strongly. Waves on ponds,

lakes, and running water. Water turbid from mud stirred up. Incaving to some extent of sand or gravel stream banks. Rang large

church bells, etc. Suspended objects made to quiver. Damage

negligible in buildings of good design and construction, slight to moderate in well-built ordinary buildings, considerable in poorly built or badly designed buildings, adobe houses, old walls (especially where

laid up without mortar), spires, etc. Cracked chimneys to considerable extent, walls to some extent. Shook down loosened

brickwork and tiles. Broke weak chimneys at the roof-line (sometimes damaging roofs). Fall of cornices from towers and high

buildings. Dislodged bricks and stones. Overturned heavy furniture,

with damage from braking. Damage considerable to concrete

irrigation ditches.

VIII Fright general, alarm approaches panic. Disturbed persons

driving motor cars. Trees shaken strongly-branches, trunks, broken

off, especially palm trees. Ejected san and mud in small amounts.

Changes: temporary, permanent; in flow of springs and wells; dry

wells renewed flow; in temperature of spring and well waters. Damage sUght in structures (brick) built especially to withstand

earthquakes. Considerable in ordinary substantial buildings, partial

collapse: racked, tumbled down, wooden houses in some cases; threw

out panel walls in frame structures, broke off decayed piling. Fall of

walls. Cracked, broke, solid stone walls seriously. Wet round to some extent, also ground on steep slopes. Twisting, fall, of chimneys,

columns, monuments, also factory stacks, towers. Moved

conspicuously, overturned, veiy heavy furniture.

IX Panic general. Cracked ground conspicuously. Damage

considerable in (masonry) structures built especially to withstand

earthquakes: threw out of plumb some wood-frame houses built especially to withstand earthquakes; great in substantial (masonry)

buildings off foundations, racked frames; serious to reservoirs; underground pipes sometimes broken.

X Cracked ground, especially when loose and wet, up to widths of

several inches; fissures up to a yard in width ran parallel to canal and

stream banks. Landslides considerable from river banks and steep

coasts. Shifted sand and mud horizontally on beaches and flat land. Changed level of water in wells. Threw water on banks of canals,

lakes, rivers, etc. Damage serious to dams, dikes, embankments.

Severe to well-built wooden structures and bridges, some destroyed.

Developed dangerous cracks in excellent brick walls. Destroyed most masonry and frame structures, also their foundations. Bent railroad

rails slightly. Tore apart, or crushed endwise, pipelines buried in earth. Open cracks and broad wavy folds in cement pavements and

asphalt road surfaces.

XI Disturbances in ground many and widespread, varying with ground

material. Broad fissures, earth slumps, and land slips in soft, wet

ground. Ejected water in large amount charged with sand and mud.

Caused sea-waves ("tidal" waves) of significant magnitude. Damage

severe to wood-frame structures, especially near shock centers. Great

to dams, dikes, embankments, often for long distances. Few, if any

(masonry), structures remained standing. Destroyed large well-built bridges by the wrecking of supporting piers, or pillars. Affected

yielding wooden bridges less. Bent railroad rails greatly, and thrust them endwise. Put pipe lines buried in earth completely out of

service.

XII Damage total-practically all works of construction damaged

greatly or destroyed. Disturbances in ground great and varied, numerous shearing cracks. Landslides, falls of rock of significant

character, slumping of river banks, etc., numerous and extensive. Wrenched loose, tore off, large rock masses. Fault slips in firm rock,

with notable horizontal and vertical offset displacements. Water

channels, surface and underground, disturbed and modified greatly.

Dammed lakes, produced waterfall, deflected rivers, etc. Waves seen on ground surfaces (actually seen, probably, in some cases). Distorted lines of sight and level. Threw objects upward in the air.

2-8
122°

Figure 2.2.1 (A) Location map of the Loma Prieta mainshock and aftershock distribution. The size

of the earthquake symbols is relative to magnitude. Black lines are faults, dashed where

inferred. (B) Cross section along the line A-B in (A) showing the mainshock hypocenter

and the vertical distribution of aftershocks. Note that the dip of the San Andreas fault

is about 70° to the southwest. (C) Schematic block diagram showing the major geologic

features in the earthquake area and coseismic deformation (from Ward and Page, 1989).

2-9
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2-10
Figure 2.3.1 Location of strong motion instrumentation. USGS stations are designated by triangles

while CDMG stations are designated by circles. The peak horizontal acceleration as a

fraction of gravity is shown adjacent to the station. The area near San Francisco and

Oakland is shown enlarged in figure 2.3.2. The records for the 10 sites indicated by the

letters A-J are shown in figure 2.3.4.

2-11

-123.00 -122.50 -122.00 -121.50 -121.00

-123.00 -122.50 -122.00 -121.50 -121.00

Figure 2.3.3 The relative peak horizontal acceleration is shown at each of the stations identified in

figure 2.3.1 by the diameter of the circles. The largest circle at site A represents an

acceleration of 0.64 g.

2-13
Site A - Corralitos (CSMIP Station 57007): 7 km from epicenter on landslide deposits

90°—-v/ViN*^-j|

Up '-WywMv^ ^

360

~^ 0.50 g

0.47 g

0.64 g

Site B - Capitola (CSMIP Station 47125): 9 km from epicenter on alluvium

360

^M]hK^^ ^ 0.47

MlTM

g

60 g

0.54 g

Site C - Santa Cruz (CSMIP Station 58135): 16 km from epicenter on limestone

90

Up—

u^wMIMMmvi^^

.6 0°—'^^--•Y.-wv\^^VV|^^

0.44 g

0.40 g

0.47 g

Site D - Lexington Dam, left abutment (CSMIP Station 57180): 19 km from epicenter on slate and limestone

E —

Up ^AA^Wify^f'^-'^^^A^^I^^VA^^,VvvV^^^

0.41 g

0.15 g 0.45 g

Site E - HoUister Airport (USGS Station 1656): 45 km from epicenter on alluvium

255 • III ^ <rfV^^«»

165- •0.27 Q

10 IS 20 Sec.

Figure 2.3.4 Ground accelerations for the stations designated A-J in figure 2.3.1. The distance from

the epicenter and the site geology are indicated.

2-14
Site F - San Francisco, Rincon Hill (CSMIP Station 58181): 95 km from epicenter on Franciscan sandstone, shale

90- ^-w^w-^ 0.09 g

Up —_ 0.03 g

360? — -— —_ -^^.^.^— ^ 0.08 g

Site G - Emeryville (USGS Station 1662): 97 km from epicenter on fill

Up — -• ^- . wy«-s,<7\^vv*-',— — . .— 0.06 g

Site H - San Francisco, Presidio (CSMIP Station 58222): 98 km from epicenter on serpentine

90^ —V— —'^/"^^^y^^V ^^-''w—.-~-v- ~ — ^ 0.21 g

Site I - Yerba Buena Island (CSMIP Station 58163): 95 km from epicenter on Franciscan sandstone

90^ ..y^ . ^— 0.06 g

Up

0.03 g

360? —-—~-—— — 0.03 g

Site J - Treasure Island (CSMIP Station 58117): 98 km from epicenter on fill

90° >^ ^.^-^^.^^^^syXl^y^ 0.16 g

Up , —— 0.02 g

360^ — — —s.,^-'~v/'^^^\XV.,j—^—— 0. 1 1 g

0 1 2 3 4 5 10 15 20 Sec.

Figure 2.3.4 (Cont'd) Ground accelerations for the stations designated A-J in figure 2.3.1. The distance

from the epicenter and the site geology are indicated.

2-15
0 10 20 30 40 60 MLOMinERS

1 I 1 I I I

Figure 2.3.5 Preliminary map showing the distribution of Modified Mercalli intensity for the October

17, 1989 Loma Prieta, California earthquake. Intensity values for localities are given in

Arabic numbers. Roman numerals represent the intensity level between isoseismal lines.

Location of the earthquake epicenter is shown by the circled star. Numbers enclosed in

circles have been added since original publication (Plafker and Galloway, 1989). Figure

2.4.1 shows more detailed assessments in the cities of San Francisco and Oakland,

California.

2-16
2-17
0 1 2 3 4 5 HLOyETERS

1 I 1 I 1 I

Damage inspections for San Francisco. The figure shows 378 inspections judged as "red"

using the ATC postearthquake safety evaluation procedures (ATC, 1989a, 1989b). Damage

judged sufficiently for a "Red" designation may be caused by either ground shaking, ground

failure, or be in a dangerous area. The preliminary intensity contour based on ground

shaking effects from figure 2.4.1 is shown for comparison.

2-18
3. SURFICIAL GEOLOGY AND FOUNDATION FAILURES

by Felix Y. Yokel

3.1 Introduction

This chapter describes observed soil displacements and failures, foundation failures, and the effect of

subsurface conditions on the severity of the observed earthquake damage.

The Loma Prieta earthquake is characterized by a selective damage pattern. While many structures in

areas of relatively shallow, competent soil deposits in the hilly sections of the Los Gatos and Santa Cruz

areas, approximately 10 miles (15 km) from the epicenter, remained undamaged or suffered only relatively

minor damage, other structures located in areas of deep Bay Mud or alluvium in San Francisco and

Oakland, over 60 miles (95 km) from the epicenter, suffered serious damage. Even though a variety of

conditions, such as the age, stiffness and lateral load resistance of the affected structures, contributed to

this selective damage pattern, it is evident from an examination of the surficial geology in areas where

major earthquake damage occurred that the subsurface conditions in these areas were a major contributing

factor.

3.2 Ground Failures

Figure 3.2.1 shows the approximate locations of observed ground failures. Four failure categories are

identified: (1) fault zone ground cracking; (2) ground failures; (3) liquefaction; and (4) tension cracks. The

observations compiled in the map were recorded by the U.S. Geological Survey in Menio Park, CA, (prior

to October 23, 1989), the University of California at Berkeley (Abolhassan et al., 1989), the U.S. Army

Corps of Engineers, and the ICSSC team.

The fault zone ground cracking locations were identified by the USGS teams. Most of the fissures are

located in the vicinity of the San Andreas and Sargent faults, within a zone extending approximately 12

miles (20 km) to the north and 18 miles (30 km) to the south from the epicenter. Some isolated cracks

were observed in the vicinity of the Calaveras fault near the Anderson and Coyote Reservoirs, and in

the vicinity of the Hayward fault near Fremont.

Ground failures, most of which were documented by the USGS teams, include landslides, tension cracks,

and observable settlements. Liquefaction was identified by telltale signs such as sand boils, ground

cracking, buckled sidewalks and lateral spreading. Some tension cracks, mostly in embankments, were

3-1
documented by the U.S. Corps of Engineers, and the ICSSC team. The failure categories identified in

figure 3.2.1 are not mutually exclusive. For instance tension cracks in embankments may be attributable

to a stability failure, in which case they would fall under "ground failures," or to liquefaction.

Most identifiable ground failures occurred within a distance of about 12 miles (20 km) fi^om the epicenter.

However, liquefaction was observed as far as 70 miles (110 km) from the epicenter (Abolhassan et al.,

1989). This is because most liquefaction phenomena occurred in areas of deep soil deposits. In these

deposits the low-frequency components of the earthquake motion were amplified (Maley et al., 1989,

Shakal et al., 1989, Plafker and Galloway, 1989). These low-frequency components decay much more slowly

with distance from the epicenter than the high-frequency components, and thus can cause damage and

trigger liquefaction at great distances from the epicenter.

3.3 Relation of Subsurface Conditions to Earthquake Damage

3.3.1 San Francisco

Figure 3.3.1 shows the surficial geology of the San Francisco area. The information was derived from

Borcherdt and Gibbs, 1975, and supplemented with information from Joyner, 1982. The boundaries

between the various soil types are only approximately drawn. Four types of subsurface conditions are

identified: Bay Mud; alluvium deeper than 90 ft (30 m); alluvium from 0 - 90 ft (0 - 30 m) deep; and

shallow bedrock.

The Bay Mud, also referred to as the "new" Bay Mud is a deposit of recent geological origin and consists

of normally consolidated (except for the surcharge effects of man-made fill and structures) organic-rich silty

marine clays with some sands with water contents of 50% or more and unconfined compression strengths

on the order of 600 psf (30 kPa) near the surface and increasing with depth at the rate of approximately

50 psf (2.4 kPa) per meter. The new Bay Mud layer is up to 130 ft (40 m) thick and has typically shearwave propagation velocities on the order of 300 - 425 ft/s (90 - 130 m/s) (Borcherdt and Gibbs, 1975).

In most on-shore locations the Bay Mud is overlain by about 20 ft (7 m) of man-made fill, which consists

of a mixture of rubble and hydraulically placed silty and clayey sands.

The alluvium comprises older Bay sediments, including "old" Bay Mud which consist mostly of stiff silty

sandy clays, and granular deposits of medium to dense silty and clayey sands, sands, and gravels. The old

Bay Mud is the oldest unconsolidated deposit (Pleistocene) and is also identified as the "Alameda

Formation" (Radbruch, 1957). The old Bay Mud was overconsolidated by desiccation during the glacial

3-2
periods and typically has water contents of less than 40%. It is therefore much denser than the new Bay

Mud. The old Bay Mud, wherever it occurs, comprises the lowest layer of alluvium and rests on bedrock.

Granular deposits, also identified as "Merritt Sands" (Radbruch, 1957) sometimes form an intervening

layer between the new and the old Bay Mud. Typical shear wave propagation velocities in the alluvial

deposits are 600 ft/s (200 m/s) at the surface and increase with depth. Thicknesses of the alluvium range

up to 1800 ft (600 m) (Borcherdt and Gibbs, 1975).

Figure 3.3.2 shows the approximate location of observed earthquake damage in the San Francisco region

together with information on subsurface conditions which was derived from the Joyner map referred to

previously. The Joyner map was prepared for the purpose of microzonation on the basis of borehole

information. The map shows two details: the boundaries of the area underlain by "Bay Mud"; and the

boundaries of the 200 ft (60 m) depth contour of surficial deposits. Within the 200 ft depth contour in

the Bay Mud area, the new Bay Mud rests on a layer of alluvium.

Two locations where major highway damage occurred, 1-280 and Evans Avenue (1), and 1-480 and Market

Street (4), are within the Bay Mud area, as well as in the 200-ft contour. An examination of available

boring logs indicates, that the 1-280 site where damage occurred, is underlain by about 10 ft (3 m) of loose

gravel fill, resting on a 40 ft (13 m) thick layer of new Bay Mud, which in turn rests on an layer of

alluvium consisting mainly of dense fine to medium sands in its upper part and stiff clays in its lower part.

Bedrock is at a depth of 190 ft (58 m). From this site the Bay Mud layer increases to a thickness of

approximately 90 ft (30 m) toward the northbound direction of 1-280.

Borings from 1-480, the Embarcadero section (4), indicate that the Bay Mud layer is between 100-120 ft

(30-37 m) thick and bedrock is at a depth of approximately 240 ft (75 m). The alluvial layer between

the new Bay Mud and the bedrock consists of old Bay Mud, overlain by a layer of granular deposits.

Highway damage location (2) on 1-280 and Channel Street where joints of bents sustained damage is also

in the Bay Mud area, but not within the 200 ft depth contour. The damage in location (2) is more

localized than that in locations (1) and (4). One of the highway damage areas shown, U.S. Rt. 101 and

Fell Street (6), is near the edge of the 200 ft contour and within the area of alluvial deposits, but not in

the Bay Mud area.

Extensive liquefaction damage was evident in the Marina District within the boundaries of the crosshatched area at location (8) in figure 3.3.2. The most intensive damage was concentrated in an area which

was filled between 1913 and 1915 in preparation for a 1915 exposition. The fill consists of approximately

3-3
30 ft (10 m) of hydraulically placed loose sand (Standard Penetration Test blow counts of 1 to 6), and is

underlain by approximately 60 ft (20 m) of new Bay Mud and another 60 ft of dense sand (alluvium).

Microseismic measurements in the Marina District reported by Watabe (1989), indicate a maximum site

period of 1.2 s.

The most prominent telltale sign of liquefaction are sand boils. These occur as a result of piping, caused

by hydrostatic excess pressures. A typical sand boil is shown in figure 3.3.3. Note that the sand was

washed out through the joint between the sidewalk tile and the brick wall retaining the stair case. Similar

sand boils were observed in many other locations throughout the Marina District. While some of the sand

probably was washed out by ruptured water mains, there are many locations, some within buildings, where

the sand boils could have been caused only by the presence of excess pore water pressures resulting from

the ground shaking.

Typical liquefaction damages observed were buckled sidewalks (fig. 3.3.4) ruptured pipelines, and tension

cracks (fig. 3.3.5). The pipeline ruptures, which occurred in more than 100 locations in the Marina

District, obviously were caused by excessive ground displacements, associated with liquefaction or cyclic

mobility. In some locations there was heave in the pavement on top of pipelines, which either did not

settle with the surrounding soil or were uplifted by buoyancy (fig. 3.3.6).

Extensive structural damage to buildings coincided with the liquefaction area in the Marina District. The

structural damage in the Marina District is discussed in detail in Chapter 4 of this report. Further studies

are needed to determine the mechanism by which liquefaction caused this damage or contributed to it.

While some settlement and tilting of foundations was observed by others (Abolhassan et al., 1989) the

structural failure mode observed by the ICSSC team was shear failure in the lowest story. The residential

structures that failed in the Marina District are for the most part three or four stories high. These

structures, under normal conditions, have resonant frequencies on the order of 3 Hz, while the

characteristic frequency of the ground motion is less than 1 Hz. Thus resonance between the ground

motion and the dynamic response of these building normally would not contribute to the initiation of

structural failures. However, in this instance, loss of foundation stiffness could have modified the dynamic

response characteristics of the structures.

Other areas underlain by Bay Mud deposits where liquefaction was observed and where a great number

of buildings had to be condemned because of structural failures are on Market Street near 1st Street (11),

on Folsom Street between 4th and 10th street (5), on S. Van Ness Ave. near U.S. Rt. 101 (10), and on

Bluxome Street (3).

3-4
Damage location (9) and other building damage locations, which are shown in the figure but not identified

by numbers, and which are remote from Bay Mud deposits, are generally located in the deep alluvial

deposit area shown in figure 3.3.1. In the case of damage location (9), the damaged buildings were

reported to be located in a man-made fill area.

In summary, most of the damages to structures and lifeline systems in San Francisco occurred in areas

of Bay Mud deposits. The most likely reason for the damages are the amplification of the ground motion

by the Bay Mud and the underlying alluvial deposits and soil liquefaction. The amplification of the ground

motion also contributed to the widespread liquefaction of fill areas. In many areas, liquefaction of the

man-made fill covering the Bay Mud deposits coincided with major structural damage to buildings and

utilities (damage locations (3), (4), (5), (8), and (10) in fig. 3.3.2).

3.3.2 Oakland

Figure 3.3.7 shows the surficial geology of the areas in Oakland where damage occurred. Three types of

subsurface conditions are identified in the figure: (1) Bay Mud, which is similar to the new Bay Mud area

identified in figure 3.3.1; (2) alluvium; and (3) shallow bedrock. The information in figure 3.3.7 is taken

from Gibbs and Borcherdt (1974), and from a map prepared by Radbruch (1957). In the figure, the

alluvial deposits are not differentiated by their characteristics. However Radbruch's map shows two types

of surficial alluvial deposits: fine silty and clayey sands with lenses of sandy clay and clay, identified as

"Merritt Sands"; and clayey gravels, sandy and silty clays, and sand-clay-silt mixtures with pebbles, identified

as the "Temescal Formation." These surficial alluvial deposits, as well as the Bay Mud deposits, rest on

a deep layer of old Bay Muds, identified by Radbruch as the "Alameda Formation." The upper portion

of this layer has been overconsolidated by desiccation during the glacial periods. Most available boring

logs, including those taken in conjunction with the construction of 1-880, do not penetrate to bedrock.

However, a boring on the Naval Reserve to the south of the inner Harbor (Boring 88 in Radbruch's map)

indicates a depth to bedrock of 440 ft (135 m), and another boring at Adeline Street in Oakland, not far

from the collapsed 1-880 viaduct, indicates a depth of 550 ft (167 m) to bedrock (Radbruch, 1957).

The most prominent failure in the Oakland area is collapse of the 1-880 viaduct which is discussed in

Chapter 5. As indicated in figure 3.3.7, the collapsed part of the viaduct is located near the outer

boundary of the Bay Mud area. An examination of available boring logs indicates that near the north

end of the collapsed area there is 5 ft (1.5 m) fill, 6 ft (2 m) of Bay Mud and 60 ft (17 m) of alluvium,

which consists of stiff silty clays (Temescal Formation). The borings stopped in a compact layer of coarse

3-5
sand and gravel and were not carried to bedrock. Toward the south end of the collapsed area the

thickness of the Bay Mud layer increases to 25 ft (8 m) and there is a layer of dense sands and gravels

(Merritt Sand) to a depth of 100 ft (32 m). All borings taken stopped in a layer of compact sands and

gravels and did not penetrate into the underlying layer of old Bay Mud. However the information

previously discussed indicates that the depth to bedrock is very great.

The area of extensive building damage in Oakland, identified in figure 3.3.7 is in the surficial alluvial

deposit identified by Radbruch as Merritt Sand and is in the vicinity of the Bay Mud deposits surrounding

Lake Merritt. Radbruch's information indicates that this area is also underlain by a deep layer of the

Alameda Formation.

It is of interest that three strong motion stations in the Oakland area, shown in figure 2.3.2 recorded peak

accelerations of 0.26g, 0.26g and 0.29g. These stations provide a measure of the amplification of the

earthquake motion attributable to the deep alluvial deposits in the area, and also indicate that over this

area of deep alluvial deposits the strong motion did not vary substantially. The strong motion records

indicate dominant site periods from 1 to 1.5 s (Maley et al., 1989).

3.3.3 Watsonville

The damaged area in Watsonville is located in the deep alluvial soil deposits of the Pajaro River valley.

The subsurface conditions in the Watsonville area have been extensively studied (Dupr6 and Tinsley, 1979,

and Dupr6, 1975). The building damage in Watsonville is in a formation of fluvial, estuarine and alluvial

fan deposits designated by Dupr6 and Tinsley as "older floodplain deposits." These deposits are

characterized as "unconsolidated, relatively finegrained deposits of sand and silt, commonly including

relatively thin layers of clay." In accordance with the information utilized by Dupr6 these deposits are

coarser grained at the bottom and become more finegrained toward the surface, and their depth is on the

order of 220 ft (67 m) or more. The surface deposits near the Pajaro River where fissures were observed

are designated as "younger floodplain deposits." These deposits are similar to the older floodplain deposits,

but they are more susceptible to liquefaction.

Extensive tension cracking was observed along the embankments of the Pajaro River (fig. 3.2.1). Figure

3.3.8 shows a 2 ft (600 mm) vdde tension crack near the bridge of Main Street over the Pajaro River.

The crack was parallel to the river between the dike and the river. There was a second parallel crack

nearer the river which was approximately 6 in (150 mm) wide. There were also some tension cracks

outside the embankment and parallel to it (fig. 3.3.9). The tension cracks in the soil parallel to the

3-6
embankment are attributable to spreading of the embankment, probably caused by liquefaction. The bridge

itself had to be closed to traffic because of structural damage. The bridge which was built in 1947 is

supported by 40 ft (12 m) long untreated wood piles. Preliminary survey information from the City of

Watsonville indicates that the bridge and a manhole in its vicinity settled approximately 6 in (150 mm).

The Pajaro River bridge is located in the area designated by Dupr6 and Tinsley as younger floodplain

deposits. '

Extensive structural and pipeline damage was observed in dovmtown Watsonville in the areas of the older

floodplain deposits. Shallow pits excavated for utility pipe repair contained deposits of fine sands,

interbedded witl^i clays (fig. 3.3.10). The observation made in these pits corroborates the observations

recorded by Dupr6 and Tinsley. No telltale signs of liquefaction were observed in the damage area in

downtown Watsonville and no groundwater was observed in the excavated pits (about 7 ft (2 m) deep).

Borings taken in March 1966 at the fire station in the downtown area show 2-3 ft (0.6 m) of silt, underlain

by a 5 ft (1.5 m) layer of clay, which in turn rests on sand. The deepest boring terminated in sandy soils

at a depth of 28 ft (8.5 m). The groundwater level at the time of boring was 17.5 - 21.5 ft (5.3 - 6.5 m)

below surface. A local building official mentioned that 1989 was a dry year prior to the earthquake.

Thus, it is unlikely that the ground water level was much closer to the surface during the earthquake to

cause liquefaction near the base of the footings in the downtown area of Watsonville. Typically, the

buildings in downtown Watsonville are supported by shallow spread footings.

3.3.4 Santa Cruz

Extensive building damage occurred at the Pacific Garden shopping mall which parallels the San Lorenzo

river. No specific subsurface information has been obtained to date, however the damaged area is located

on the alluvial deposits of the San Lorenzo River valley, which are similar to the "older and younger

floodplain deposits" previously described (Dupr6, 1975). Not much damage was observed in the hilly

sections of Santa Cruz. Microseismic measurements in the Pacific Garden mall area (Watabe, 1989)

indicate a maximum site period of 3.26 s, which would be associated with deep soil deposits.

Many of the damaged buildings were stiff masonry shear wall structures which are judged to have natural

frequencies much higher than the characteristic frequency of the site. While widespread liquefaction was

observed by others in the Santa Cruz area (i.e. Abolhassan et al., 1989), no telltale signs of liquefaction

were observed by the ICSSC team in the Pacific Garden mall area.

3-7
3.3.5 Los Gatos

Most of the earthquake damage in Los Gatos occurred in the flat terrain adjacent to the Los Gatos Creek.

This is also the area where most of the unreinforced brick structures are located. The subsurface

information obtained on the site is from shallow pits dug to repair the extensive damage to utilities. The

soils in these pits were granular (sands and gravels) and stratified, indicating that these are alluvial deposits

(fig. 3.3.11). No groundwater accumulated in the 7 ft (2 m) deep pits.

The Los Gatos area was mapped by Dibblee and Brabb (1978). In accordance with their map, most of

the damage occurred in an area of surficial deposits of recent (Holocene) alluvium, resting on older,

Pleistocene alluvial; deposits. No information was obtain^ed on the depth of these deposits.

3.4 Summary ,

The pattern of earthquake damage was closely correlated with subsurface conditions. Most structural and

lifeline damage occurred in areas of deep soil deposits. Heavy concentrations of damage to structures and

utilities were associated with soil liquefaction.

In San Francisco most of the damage occurred in areas underlain by new Bay Mud, but there also were

damages in areas of deep alluvial deposits. The damage in Oakland covered in this report occurred in an

area of deep alluvial deposits, which exhibited approximately uniform amplification of the ground motion

over a large region. At the collapsed 1-880 viaduct these alluvial deposits were covered with a 6-24 ft (2-

8 m) thick layer of Bay Mud. It is not yet known whether this soft layer contributed to the collapse. In

Watsonville and Santa Cruz most of the earthquake damage occurred in areas of deep alluvium. In Los

Gatos most of the damage also is in an alluvial area, however there is no information on the depth of

the deposit.

The effect of the subsurface conditions on the damage pattern observed in this earthquake is mostly

attributed to amplification of the earthquake motion by the underlying soil deposits (Plafker and Galloway,

1989). When maximum accelerations from rock sites of roughly comparable epicentral distance are

compared with those from sites of new Bay Mud deposits and sites of alluvial deposits the average

amplification ratios observed in this earthquake are 1.8 and 2.6 for vertical and horizontal accelerations,

respectively, for Bay Mud sites, and 1.9 and 1.8 for vertical and horizontal accelerations, respectively, for

alluvial sites (EERI, 1989). These amplifications of maximum accelerations are associated with much

greater amplifications of velocities and displacements in the horizontal direction, because of the long

3-8
characteristic period of the horizontal components of the earthquake motion in the deep soil deposits.

The long characteristic period of the deep soil deposits also is a reason why severe damage occurred in

areas with epicentral distances of approximately 60 miles (95 km). Low-frequency components of the

earthquake motion decay more slowly with epicentral distance than high-frequency components and can

be strongly amplified by deep soil deposits. The probability of experiencing severe shaking is increased

in areas of deep soil deposits because strong shaking can occur from distant as well as nearby earthquakes.

Strong correlations between subsurface conditions and damage patterns have been observed in many other

earthquakes. For instance, in the September 19, 1985 Mexico City earthquake most of the structural

damage occurred in areas of the lakebed area where the depth of the soft clay deposits ranged from 100 -

125 ft (30 - 38 m) (Stone et al., 1987). However, in the Mexico City earthquake, most of the damage

was caused by resonance between the ground motion and the dynamic response of the damaged buildings.

In the Loma Prieta earthquake, relatively stiff buildings also were damaged in areas of deep soil deposits,

while similar buildings in areas of shallow, competent soil deposits remained undamaged. The vulnerability

of these latter structures cannot be attributed to amplification resulting fi^om resonance between the ground

motion and the initial dynamic response of the structures.

At present, design criteria recognize the effect of site characteristics on structures having long natural

periods. However, for structures with natural periods less than approximately 0.6 s no site amplification

is recognized, and for soil profiles containing soft clay, recommended design spectra actually stipulate a

reduced seismic design coefficient (FEMA, 1988, NEHRP Section 4.2.1). These provisions and those for

retrofit of existing structures should be re-examined.

3.5 References

Abolhassan et al. 1989, Preliminary Report on Seismological and Engineering Aspects of the October 17,

1989 Santa Cruz (Loma Prieta) Earthquake, University of California at Berkeley, Report No. UBC/EERI89/14.

Borcherdt, R. D. and Gibbs, J. F., 1975, "Prediction of Maximum Earthquake Intensities for the San

Francisco Bay Region," U.S. Geological Survey Open File Report 75-180.

Federal Emergency Management Agency, 1988, "NEHRP Recommended Provisions for the Development

of Seismic Regulations for New Buildings," FEMA 95.

California Division of Mines and Geology, 1989, "Strong Ground Shaking from the Loma Prieta

Earthquake of October 17, 1989 and its Relation to Near-Surface Geology in the Oakland Area,"

November, 1989.

3-9
Dibblee, Jr., T. W. and Brabb, E. E., 1978, "Preliminary Geologic Maps of the Chittenden, Los Gatos, and

Watsonville East Quadrangles," U.S. Geological Survey Open File Report 78-453.

Dupr6, W. R., 1975, "Quaternary History of the Watsonville Lowlands, North-Central Monterey Bay

Region," Doctoral Dissertation, Stanford University.

Dupr6, W. R. and Tinsley III , J. C, 1975, "Maps Showing Geology and Liquefaction Potential of the

Northern Monterey and Southern Santa Cruz Counties, CA," U.S. Geological Survey Misc. Field Studies,

Map MS-648.

Earthquake Engineering Research Institute and National Research Council, 1989, "Loma Prieta Earthquake,

October 17, 1989, Preliminary Reconnaissance Report," EERI 89-03.

Gibbs, J. F. and Borcherdt, R. D., 1974, "Effects of Local Geology on Ground Motion in the San

Francisco Bay Region, A Continued Study," U.S. Geological Survey Open File Report 74-22.

Joyner, W. B., 1982, "Map Showing the 200-foot Thickness Contour of Surficial Deposits and the Landward

Limit of Bay Mud Deposits of San Francisco, CA", U.S. Geological Survey Misc. Field Studies Map MF1376, 1 Plate, Scale 1:24,000.

Maley, R., et al, 1989, "U.S. Geological Survey Strong-Motion Records from the Northern California

(Loma Prieta) Earthquake of October 17, 1989," Open-File Report 89-568, U.S. Geological Survey.

Plafker, G. P. and Galloway, J. P., Eds., 1989, "Lessons Learned from the Loma Prieta, CA, Earthquake

of October 17, 1989.," U.S. Geological Survey Circular 1045, U.S. Government Printing Office.

Radbruch, D. H., 1975, "Areal and Engineering Geology of the Oakland West Quadrangle, CA," U.S.

Geological Survey Misc. Geologic Investigations, Map 1-239.

Shakal, A., et al., 1989, "November 1989 CSMIP Strong Motion Records from the Santa Cruz (Loma

Prieta) CA Earthquake of 17 October 1989," California Division of Mines and Geology, Report OSMS

89-06.

Stone, W. C, Yokel, F. Y., Celebi, M., Hanks, T., and Leyendecker, E. V., 1987, "Engineering Aspect of

the September 19, 1985 Mexico Earthquake," NBS Building Science Series 165, National Bureau of

Standards.

Watabe, M., 1989, "San Francisco Earthquake," Tokyo Metropolitan University.

3-10
HAYWARD

FAULT

San Franaij

Pacific

Ocean

(17) SAN

ANDREAS

FAULT

Los Gatos

• Ground Failure

o Fault Zone Ground Cracking

X Sand Boils (Liquefaction)

Tension Crack

® Epicenter

Santa Cruz w^soriv(iie

8<

x

Monterey

CALAVERAS

FAULT

.Anderson Lake

Coyote Lake

SARGENT

FAULT

Figure 3.2.1 Ground failures caused by the Loma Prieta earthquake.

3-11
Outer limit of bay mud deposits iBay mud

Coutour of 60m (200 ft) depth of ^^^^^ ^^^^^ ^'^'^^

unconsolidated deposits i j Alluvium, 0-30m thick

Shallow bedrock

Surficial geology of the San Francisco area.

3-12

1
Outer limit of bay mud deposits

^ — Coutour of 60m (200 ft) depth of

unconsolidated deposits

C7

^

Area of damage concentration

® Major damage to highway structures

O Minor damage to highway structures

A Major damage to buildings

A Minor damage to buildings

Figure 3.3.2 Correlation of damage patterns and surficial geology in the San Francisco area.

3-13
Figure 3.3.4 Typical liquefaction damage in the Marina District of San Francisco.

3-14
Figure 3.3.6 Pavement heave over utility pipe.

3-15
Figure 3.3.7 Surficial geology of the Oakland Area.

3-16
Figure 3.3.9 Tension cracks behind Pajaro River levee.

3-17
Figure 3.3.11 Excavated pit in Los Gatos.

3-18
4. PERFORMANCE OF BUILDINGS

by Nicholas J. Carino

4.1 Introduction

This chapter describes the performance of buildings within the Bay area and within the epicentral region.

Because of the large area in which buildings were damaged, the inspections concentrated on those regions

which experienced a large amount of severe damage. As discussed in Chapter 3 the most serious damage

was concentrated in areas of deep soil deposits.

Most structures designed according to modern codes and standards performed well without structural

damage. The majority of the damaged structures were either wood-framed dwellings or unreinforced

masonry buildings which had not been strengthened to increase their seismic resistance. It should be

noted that the Loma Prieta earthquake was not a severe test of buildings designed according to modern

seismic criteria. Except for the epicentral region, peak ground accelerations were less than implied by

modern building codes. Also, the duration of strong ground shaking was only 10 to 15 seconds. Thus,

the favorable performance of the majority of modern buildings can not be used as evidence that current

seismic design criteria are adequate.

There were isolated failures in modern structures to which the team did not gain access. Detailed

investigations of the circumstances associated with these failures can provide opportunities to improve

design standards if it is found that the failures occurred where standards were met. The Loma Prieta

earthquake also provided opportunities to evaluate the performance of the various seismic strengthening

methods used within the affected region. The team observed many instances of successful strengthening

measures, but there were also examples of unsuccessful measures.

This chapter provides examples of the type of damage observed in the more severely affected regions.

Probable failure mechanisms are discussed where possible, but no attempt is made to provide in-depth

explanations of the underlying factors leading to the observed damage.

4.2 San Francisco

The following areas in San Francisco sustained the most severe damage and were investigated: the Marina

District, the Financial District, the Civic Center District, and South of Market District. Following the

earthquake, building officials and volunteer engineers inspected over 8500 buildings. It has been reported

4-1
by the Department of Public Works^ of the city of San Francisco that about 260 buildings were given the

status "unsafe" and about 1400 were classified as "limited entry."

4.2.1 Marina District

Figure 4.2.1 is a map of the Marina District showing the approximate areas covered by aerial photographs

(numbered 1 through 4) to be shown. Figure 4.2.2 is an aerial view (#1 in fig. 4.2.1) of the northwestern

quadrant of the Marina District. This photograph gives a general impression of the type of construction

in this area. Most buildings are of wood frame construction with stucco or brick masonry veneer exteriors.

The majority of the buildings are three or four stories tall. In general, the buildings at the corners of the

blocks are four-story, apartment buildings with garages at the first story level. The mid-block buildings

tend to be three stories and are built with little, if any, clearance between them.

Figure 4.2.3 is a map showing the area (bounded by Baker Street, Marina Boulevard, Fillmore Street, and

Chestnut Street) which was secured by the police to control access to the zone of severe damage. The

table in figure 4.2.3, lists the numbers of buildings that were subsequently declared unsafe by the

Department of Public Works. While the damage was concentrated within the area shown on the map,

buildings with severe damage were scattered throughout the District.

The degree of damage ranged from total collapse to cosmetic damage to the building exteriors. Figure

4.2.4 is an aerial view (#2 in fig. 4.2.1), looking northwest, of the intersection of Fillmore Street and

Cervantes Boulevard. The heap of rubble is the remains of a collapsed four-story building, which was

subsequently demolished for safety reasons. Three residents, including a 3-month-old child, were killed

in the collapse. Figure 4.2.5 is an aerial view (#3 in fig. 4.2.1), looking toward the southeast, of the

intersection of Divisadero and Beach Streets. A building collapsed at the northeast corner of Beach and

Divisadero and was subsequently demolished for safety reasons. The dark area on the northwest corner

of Beach and Divisadero is the site of the large fire in which one person died. Other severely damaged

buildings can also be seen in the photograph.

Figure 4.2.6 shows the portion of the Marina District that was examined, and the numbers indicate the

locations of buildings highlighted in the report. Figures 4.2.7 and 4.2.8 show two examples of building

which collapsed. The two four-story buildings (location #14) shown in figure 4.2.7 were severely damaged

^Database provided by Franklin Lew, Seismic Safety Program Manager, Department of Public Works,

San Francisco.

4-2
r

by the earthquake, and 5 days later, the first stories of both buildings collapsed as shown in the

photograph. These buildings were subsequently demolished. The collapsed building shown in figure 4.2.8

(location #3) is also believed have been four-stories tall, but in this case the lower three stories collapsed.

During the collapse, the building moved a large distance toward the south.

Figures 4.2.9 through 4.2.14 are examples of buildings which suffered severe damage but did not collapse.

Examination of the building geometries and locations (figure 4.2.6) reveals that the buildings had the

following common characteristics: (1) they were four stories tall; (2) the first story contained many

openings for garages; and (3) they were located at the corners of intersecting streets.

Figure 4.2.15(a) is an elevation view of a schematic representation of one of these typical corner buildings.

The garage door openings in the first story results in buildings with low lateral stiffness and strength.

Because of their corner locations, these buildings are free to move in two perpendicular directions.

Because of these factors, horizontal movement of the foundation during an earthquake would be expected

to cause large distortions in the garage story, as illustrated in figure 4.2.15(b). Figure 4.2.16 is a closeup view of the lower story at the corner of the building shown in figure 4.2.11. The brick veneer and the

wood sheeting have been stripped form the studs, revealing the ineffective lateral bracing in the only solid

wall portion of the first story.

Failure of brick masonry veneers were common throughout the Marina District. While such failures may

not be structurally significant, falling bricks posed a life hazard during the earthquake. Figure 4.2.17 shows

a four-story building (location #2) with multiple garages on the first story. Shaking of the building did

not result in large distortions as seen in the previous examples, but it was strong enough to dislodge a

large portion of the masonry facade. The building shown in figure 4.2.18 (location #10) is also a fourstory building with garages, in which more than two stories of brick facade fell. The final example

(location #12) shown in figure 4.2.19(a) is the same type of structure, and the brick masonry veneer peeled

off at a corner (similar to failure shown in fig. 4.2.16, but less severe). Figure 4.2.19(b) is a close-up of

the failed corner showing the condition of the wooden boards behind the brick. The wood is decayed and

there is no evidence of effective ties between the wood sheeting and brick veneer. At other buildings

with masonry facade failures, it was observed that ordinary nails had been used to tie the masonry to the

building. The nails were badly corroded and provided no anchorage to the masonry.

The above examples have illustrated the severity of damage to some of the buildings in the Marina

District. According to the Department of Public Works database, about 100 buildings were "red-tagged,"

which means that they were judged to be unsafe to enter. Thus the majority of the buildings in this

4-3
district were undamaged or suffered only minor damage. For example, figure 4.2.20 shows an undamaged

four-story building on Broderick Street (location #7) opposite fi-om the severely damaged building shown

in figure 4.2.12. As can be seen in the aerial photograph (view #4) shown in figure 4.2.21, these two

buildings have about the same overall shapes. Apparently, the undamaged building had been retrofitted

and some of the garage doors were removed and replaced with solid wall, thereby increasing the lateral

strength of the first story. Another example of building (location #8) with excellent performance is shown

in figure 4.2.22 (a). This three-story building had been retrofitted by using a moment resisting steel frame

as shown in figure 4.2.22(b). The only damage to the building was a broken pane of glass, although the

four-story building on the right (see fig. 4.2.22 (a)) was leaning on it.

In summary, dwellings in the Marina District suffered varying degrees of damage. The severe damage

tended to be concentrated within two blocks from the intersection of Beach and Divisadero Streets. There

was evidence of liquefaction and large soil displacements within the damage zone, but damages were

generally consistent with strong ground shaking rather than foundation failures. The most severe damage

occurred to four-story apartment buildings with garages in the first story and located at street corners.

Brick veneers collapsed due to inadequate anchorage to the wood framing.

4.2.2 Financial, South of Market, and Civic Center Districts

Figure 4.2.23 shows the locations of the other districts in San Francisco that were investigated. Also

shown in the figure are those areas which have a high concentration of unreinforced masonry (URM)

buildings. As can be seen the three districts have a large number of URM structures. As was previously

discussed, another important feature of these districts is that they include areas underlain by deep soft soil

deposits, which amplified the ground motion. The combination of seismically vulnerable URM buildings

and large ground movements resulted in extensive damage within these districts. As of October 28, the

South of Market District was reported to have had 30 structures posted as "unsafe," the second highest

number in a San Francisco district^

Figure 4.2.24 shows the locations of structures discussed in this section. As shown in figure 4.2.23,

Chinatown contains many URM buildings. Because these buildings are founded on firm ground, they were

subjected to small ground movements and none of them were damaged as severely as the collapsed wooden

structures in the Marina District. However, significant cracking was observed in some buildings. Figure

^San Francisco Examiner, October 28, 1989

4-4
4.2.25 shows a two-story building (location #1) with severe diagonal cracking in the masonry piers between

the first-story windows.

The most severe damage observed in the Financial District was concentrated in the circled region shown

in figure 4.2.24. There were indications of large ground displacements within this zone. Figure 4.2.26(a)

is a view along Davis Street, between Sacramento and California Streets (location #2), which shows

distortions in the sidewalk. The close-up view in figure 4.2.26(b) shows that there is about 1 in. (25 mm)

of permanent settlement of the sidewalk relative to the retaining wall. The horizontal scrape mark on the

wall suggests that the sidewalk moved upwards about 3 in. (75 mm) during the earthquake. The sidewalk

was adjacent to a building of modem construction which had no signs of damage, even though it had many

windows in the ground story.

Two historic URM structures on Front Street experienced severe structural damage. Figure 4.2.27(a) shows

the six-story Marine Building (location #3) on the northeast corner of Front and California Streets. While

there does not appear to be much damage, the close-up view of the northwestern corner in figure 4.2.27(b)

shows severe cracking, which starts at a level coincident with the roof line of the adjacent two-story

structure. Thus pounding between the two buildings may have contributed to the damage of the Marine

Building. The damage was so extensive that the building was subsequently demolished.

Figure 4.2.28(a) shows the structural damage suffered by the Golden State Bank building (location #4).

The southeast corner of the third story was destroyed along with the lintel. The northeast corner was also

severely damaged as seen in figure 4.2.28(b). There also appears to have been pounding between this

building and the adjacent building (location of Harrington's Restaurant).

Figure 4.2.29(a) shows the historic, four-story DeBernardi warehouse building on the corner of Front and

Sacramento Streets (location #5). This building is adjacent to Harrington's Restaurant shown in the

previous photograph. In 1972, the structure was seismically upgraded by the addition of the X-bracing

which can be seen through the windows. Figure 4.2.29(b) is a close-up view of one of the braces showing

me damage, which indicates that they were subjected to large loads during the earthquake. Thus the

retrofit measure performed its intended function of protecting the structure.

A final example of the type of damage within the Financial District is shown in figure 4.2.30. The Bank

of California building, at the corner of California and Sansome Streets (location #6), suffered major

cracking of its stone facade.

4-5
Figure 4.2.31 is an aerial view of a portion of the South of Market District. The view is toward the

southwest and the elevated highway is the ramp structure connecting the Embarcadero Freeway and the

San Francisco-Oakland Bay Bridge. As was shown in figure 4.2.23, there are many URM buildings in this

district, and a large number are used commercially. As a result, besides suffering loss of property, this

district sustained a great loss in business activity.

During the drive through the South of Market District, damaged URM buildings were observed. In

addition, there was damage to engineered structures with masonry facades. An example is the eight-story

Wells Fargo Bank Building shown in figure 4.2.32 (location #7), which suffered extensive cracking to the

masonry cladding. In 1969, San Francisco enacted the Parapet Ordinance requiring the strengthening of

parapets to resist seismic loads. As a result, a majority of the parapets have been braced and survived

the shaking.

Damage was observed in the reinforced concrete building at the corner of Fremont and Howard Streets

(location #6) shown in figure 4.2.33(a). This building can be seen at the bottom right-hand corner of the

aerial photograph (fig. 4.2.31). The building is of flat-slab construction (with column capitals) and shear

walls around the stairwell. It originally served as a warehouse, but was converted to an office building.

Extensive damage occurred to most of the spandrels on the side of the building facing Howard Street.

The cracking tended to be horizontal, except near the corners were X-cracks were formed as seen in figure

4.2.33(b).

The greatest loss of life due to a building failure occurred at a large URM structure on Bluxome Street

(location #8). The structure consisted of two similar wings. Figure 4.2.34(a) shows the appearance of

the middle of the structure where the two wings joined. The fourth-story wall of the wing to the right

collapsed into the courtyard area. The building had undergone previous strengthening as indicated by the

large bolts on the facade. These bolts went through the building, tying together the front and rear walls.

However, similar bolts were not used in the other direction, and as a result wall failures occurred.

Figure 4.2.34(b) shows the exterior wall facing Sixth Street. The entire fourth-story exterior wall collapsed.

Unfortunately, people were driving their cars along the street, and five persons were crushed to death by

the falling masonry. Reports in the local newspapers indicated that this building had undergone

engineering evaluations as early as 1982, and it had been identified as being seismically deficient. In March

of 1986, bricks were reported to have fallen from the building causing damage to a car.

The Civic Center area was the other location surveyed by the team. Damage was observed to the masonry

exterior of the Public Library, a steel-framed building. However, according to subsequent reports, the

4-6
structural frame was not damaged. Some cosmetic damage was also observed in the interior of City Hall.

The team inspected the interior of the Old Federal Building (location #9) and noted damage to interior

partitions. The most serious damage in the Old Federal Building occurred to the elevators. As shown

in figure 4.2.35, the counterweight jumped out its track during the shaking and collided with the upward

moving cab.

In summary, unreinforced masonry buildings, masonry infilled walls, and masonry cladding sustained damage

in the Financial, South of Market, and City Center Districts. A contributing factor appears to be the

amplification of ground motion by the deep soil deposits in these districts. Buildings on firmer ground

sustained little or no damage. Parapets braced and anchored to the roof framing performed well.

4-7
Figure 4.2.1 Map of Marina District showing approximate areas covered by aerial photographs.

Figure 4.2.2 Aerial view (#1) of the western portion of the Marina District, view is toward the north.

4-8
I

I

Limited access zone

Approximate Number of Buildings Declared Unsafe

Alhambra 2 Fillmore 4

Avila 10 Francisco 7

Bay St. 13 Jefferson 10

Beach 8 Marina 3

Broderick 2 Northpoint 7

Capra 1 Prado 2

Cervantes 11 Scott 18

Divisadero 12

Source: San Francisco Department of Public Works

Figure 4.2.3 Map of Marina District showing region of concentrated damage.

4-9
Figure 4.2.4 Aerial view (#2) of demolished building at the intersection of Fillmore St. and Cervantes

Blvd.

Figure 4.2.5 Aerial view (#3) of portion of the Marina District which suffered extensive damage; dark zone is location of the fire.

4-10
^ = Fire — — Inspection Route »»e««e««i«<s««««»e« Limited access zone

Figure 4.2.6 Map of the Marina District showing locations of buildings mentioned in this report.

Figure 4.2.7 Two four-story buildings which suffered total collapse of their first stories (location #14).

4-11
Figure 4.2.8 Four-story building which suffered total collapse of its lower stories (location #3).

Figure 4.2.9 Example of four-story building with multiple garages in the first story (location #1).

4-12
Figure 4.2.11 Example of four-story building with multiple garages in the first story (location #5).

4-13
Figure 4.2.13 Example of four-story building with multiple garages in the first story (location #9).

4-14
Figure 4.2.14 Example of four-story building with multiple garages in the first story (location #13).

(a)

Garage

Doors

(b)

Ground Movement

Figure 4.2.15 (a) Schematic elevation view of a four-story building with multiple garages and (b) expected

building distortion due to ground motion.

4-15
Figure 4.2.16 Detail of ineffective framing system in corner of building at the intersection of Beach and

Divisadero Streets (location #5).

4-16
Figure 4.2.17 Example of building with failure of brick masonry facade (location #2).

Figure 4.2.18 Example of building with failure of brick masonry facade (location #10).

4-17
Figure 4.2.19 (a) Example of building with failure of brick masonry facade (location #12) and (b) closeup view of corner showing damaged wood and lack of anchorage to the masonry.

4-18
Figure 4.2.21 Aerial view (#4) of the buildings shown in figures 4.2.12 and 4.2.20.

4-19
Figure 4.2.22 (a) An undamaged, three-story building (location #8) and (b) view of the moment-resting

frame used for retrofitting.

4-20
-.^.y-:.-^.y--y--.%:t:--'':y:-y>J^--

:.y-:-y-:y--y--y--:w>Jt-.-^:yyy: y -y

.^j^-y^-y::

^ ' '•y^y':-

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y'*'y^^*\y^' '"^v"^ .^^^.'••^•^•^s.c--':^ y \.y*-y''y^y '-y -y^-y '-^^^y '^y-.y-.y-.y-.y'yy^yZ--y^.

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SourceiSan Francisco Examiner

More than 11 URM buildings per block

6 to 10 URM buildings per block

District Boundary

Figure 4.2.23 Map showing location of unreinforced masonry buildings within the Financial, South of

Market, and Civic Center Districts.

4-21
Figure 4.2.24 Location of structures within the Financial, South of Market, and Civic Center Districts

discussed in the report.

4-22
Figure 4.2.25 Example of damage to unreinforced masonry building in the Chinatown area (location #1).

4-23

I
Figure 4.2.26 (a) Sidewalk showing distortions due to differential settlements, and (b) close-up view of

intersection of sidewalk with retaining wall showing magnitude of differential movement.

4-24
Figure 4.2.27 (a) Severely damaged Marine Building at corner of Front St. and California St.; (b) Closeup view of the northwest corner showing severity of cracking.

4-25
Figure 4.2.28 (a) Damage to the southeastern corner of the Golden State Bank Building; (b) Closeup view of the northeastern corner; damage was likely caused by pounding against the

adjacent building.

4-26
Figure 4.2.29 (a) The historic DeBernardi warehouse building showing the bracing added during seismic

retrofitting; (b) close-up view of the bracing showing damage caused by earthquake.

4-27
Figure 4.2.30 Damage to the stone facing of the Bank of California building.

4-28
Figure 4.2.31 Aerial view of the south of Market District, looking toward the southwest.

Figure 4.2.32 Example of damage to masonry cladding of Wells Fargo Bank building (location #7).

4-29
Figure 4.2.33 (a) Reinforced concrete, flat-slab building which sustained severe damage to the exterior

walls (location #6); (b) close-up view of damage to spandrels.

4-30
Figure 4.2.34 (a) Damage to upper story of unreinforced masonry building on Bluxome St. (location #8);

(b) damage to the southwestern wall which resulted in the death of six persons.

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4.3 City of Oakland

According to newspaper reports^ the city of Oakland suffered about $1.3 billion in damage. Two days

after the earthquake it was estimated'' that 1300 homes had been damaged, 141 of them seriously and 10

had to be demolished. Over 100 commercial structures were reported damaged, nine of which were total

losses. In addition, four public buildings were closed, including City Hall.

Figure 4.3.1 is map of downtown Oakland which includes the area observed during the inspection.

According to newspaper reports, at least five seriously damaged buildings were within the vicinity of 20th

Street and Broadway.

City Hall was reported to have sustained extensive damage, however, no damage was observed to the

exterior stone facade. Apparently, the damage occurred primarily to the interior, and it was reported that

about $30 million would be needed to restore and upgrade the structure.

The damaged observed in the area inspected was confined primarily to steel-frame buildings with brick

masonry cladding and to URM buildings. The Dalziel apartment building (location #1) shown in figure

4.3.2(a) had extensive cracking to its masonry facade. As can be seen figure 4.3.2(b), numerous diagonal

cracks formed in the piers between the windows. The Oakland Hotel (location #6) shown in figure

4.3.3(a), which served as low-income housing, was also severely damaged. It was reported' that this

building had been retrofitted by anchoring the masonry walls to the steel frame so that they would not

fall in the event of an earthquake. As can be seen in the detailed photograph shown in figure 4.3.3(b),

there was severe cracking of the masonry infill in the west wing and some bricks fell. The east wing of

the building was not observed to have the same severity of cracking. Apparently the structural frame was

not damaged as it is reported that the building is still occupied. A reinforced concrete block masonry

building being constructed opposite the Oakland Hotel (see fig. 4.3.3(a)) was not damaged.

Figure 4.3.4 shows the three-story Clay Building (location #2) which had a serious failure of its

unreinforced masonry walls. On the corner opposite to the Clay Building there was a two-story URM

'San Francisco Examiner, October 24, 1989

^e Tribune (Oakland), October 19, 1989

'Stephen Mahin, presentation at EERI-NRC-NCEER Congressional Briefing on the Loma Prieia

Earthquake, Nov. 30, 1989, Washington, D.C.

4-33
building (location #3) which was apparently built as a produce market in 1917. As is shown in figure

4.3.5, there was failure of the parapet as well as a portion of the infilled wall. The building occupied the

entire block, and the eastern portion, which was being renovated, did not appear to have been damaged.

Another example of a URM-building failure (location #5) is shown in figure 4.3.6. The collapse of the

outer wall crushed three vehicles parked in the adjacent lot. The high-rise steel frame structure, seen in

the background of figure 4.3.6, being constructed at 12th Street and Broadway, suffered damage at the top.

It was reported* that some steel members and the construction elevator fell to the ground. Fortunately

there were only some minor injuries from what could have been a serious accident.

Along Broadway and Franklin Streets, several major unreinforced masonry buildings sustained severe

damage. There was damage to infill walls of steel-framed buildings, and there was spalling of masonry

veneers due to pounding of buildings.

Damage also occurred to some modern buildings. Figure 4.3.7 shows the Trans Pacific Centre at 11th

Street and Broadway (location #4). The building had to be evacuated not because of structural damage,

but because of ruptured sprinkler lines, broken windows, and fallen ceiling tiles. The 16-story

telecommunications building at 20th and Franklin Streets (location #7) suffered damage. The structural

system consists of a steel frame with reinforced concrete shear walls. As is shown in figure 4.3.8, extensive

damage occurred to the shear walls as well as other concrete structural members.

In summary, the downtown area of Oakland suffered significant damage. Steel-framed buildings with

masonry cladding and unreinforced masonry buildings were affected. In addition, damage was sustained

by relatively new buildings. It will be of value to study why these new buildings were damaged, so that

it can be determined whether changes to design practices are needed.

^an Francisco Examiner, October 23, 1989

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Inspection route

Figure 4.3.1 Downtown Oakland showing inspection route and buildings referenced in report.

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Figure 4.3.2 Dalziel apartment building: (a) overall appearance and (b) close-up view of cracking in

masonry piers (location #1).

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Figure 4.3.4 Damage to the Clay Building, an unreinforced masonry structure (location #2).

Figure 4.3.5 Extensive damage occurred to a two-story unreinforced masonry building built in 1917

(location #3).

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Figure 4.3.6 Collapse of unreinforced masonry wall crushed vehicles in adjacent parking lot (location

#5). In background is a steel-framed building under construction which sustained damage.

Figure 4.3.7 Trans Pacific Centre Building which suffered extensive damage due to ruptured water pipes,

broken windows, and fallen ceiling tile (location #4).

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4.4 Epicentral Region

The investigative team visited Los Gatos, Watsonville, and Santa Cruz, Hollister, and Moss Landing (see

fig. 4.4.1). These communities, except for Hollister, are located within a 30-mile (50-km) radius of the

epicenter and were subjected to strong ground shaking. Strong motion records indicate the peak horizontal

and vertical accelerations ranged from 0.4 to 0.5 g, as compared to 0.64 g at the epicenter. The majority

of structures in this region are unreinforced masonry commercial buildings ranging from one to three

stories in height and wood-framed dwellings. The patterns of observed damage were similar to what has

been observed in previous earthquakes having similar ground shaking intensities. These include: collapse

of masonry chimneys at the roof line, houses sheared off of their foundations, and partial or total collapse

of unreinforced masonry walls.

In Los Gatos, about 400 out of 11,000 houses were damaged. Figures 4.4.2 (a) and (b) show examples

of damage to wood-framed houses which shifted off of their foundations. The shifting was usually as a

result of the failure of the supporting perimeter walls. Figure 4.4.3 illustrates how a house, such as those

which failed, is supported. The concrete footings support a short perimeter wall, known as a "cripple stud

wall," which supports the frame of the house. The cripple stud walls are typically built using 2-by-4 studs

and wooden boards as sheeting. When the ground moved during the earthquake, the cripple stud walls

were unable to resist the induced horizontal forces. This caused the walls to sway until they could no

longer support the weight of the house. The house crashed to the ground and secondary damage resulted

from the impact. The same type of damage was also observed to houses in Watsonville.

The communities in the epicentral area have a concentration of old URM buildings in their central

commercial districts. Extensive damage occurred in these areas. There were numerous parapet failures,

which in many cases caused extensive damage to adjoining buildings. In addition, many buildings suffered

extreme damage due to inadequate lateral resistance as illustrated by the building in figure 4.4.4 (a). With

many openings in the first story, masonry-clad, wood-framed structures, such as this, swayed excessively

causing inadequately anchored brick veneers to fall. Along Main Street in Watsonville, many of the

storefront windows were destroyed as shown in figure 4.4.4(b). The large openings in the first story of

most buildings lead to low lateral strength and the excessive deformations during the shaking lead to

glass failures.

Several historic masonry structures were severely damaged in Watsonville. The building shown in figure

4.4.5, built in 1893, had extensive failure of the parapet. The falling brick killed a woman in front of the

bakery shown on the right side of the photo. Historic St. Patrick's Church, shown in figure 4.4.6, was also

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damaged. The church, constructed in 1903, had major failures of the exterior walls, and the steeple

developed several large diagonal cracks. The hazard posed by the damaged steeple caused local officials

to block off the northern end of Main Street leading to the business district. It was observed that

retrofitted or modern structures at the northern end of Main Street were not damaged.

In Santa Cruz, the most concentrated damage to commercial structures occurred in the Pacific Garden

Mall. About one-third of the buildings suffered severe damage. Figure 4.4.7 is a map of downtown Santa

Cruz showing the portion of Pacific Avenue comprising the outdoor shopping mall. Aerial photographs

of the mall area, which were taken on October 19, 1989, are shown in figure 4.4.8. The approximate

coverage of the photographs is indicated on the map in figure 4.4.7.

At the center of figure 4.4.8(a) there is a two-story masonry structure at the comer of Pacific Avenue and

Water Street (location #1). Figure 4.4.9 shows the appearance of the structure at street level. It was

designed about 15 years ago according to the 1973 Uniform Building Code, and it survived without any

sign of structural damage. Adjacent to this modern building is an older unreinforced masonry building,

which is seen on the left side of figure 4.4.9. Although the facade shows a small amount of damage, the

roof of the building collapsed and killed two persons.

Another roof collapse in the mall caused the death of one person. This failure occurred to the

department store on the corner of Pacific Avenue and Cathcart Street, and can be seen in figure 4.4.8(c).

The roof failure was caused by the collapse of the top story masonry wall of the adjacent building. This

type of damage was observed at several other stores within the mall.

One of the historic structures severely damaged is the Cooper House located on the corner of Pacific

Avenue and Cooper Streets (location #2). The unreinforced masonry building was built in the 1880's as

the county courthouse. The building was reported to have survived the 1906 San Francisco earthquake',

but it was extensively damaged during the Loma Prieta earthquake. As shown in figure 4.4.10, portions

of exterior walls collapsed and there were extensive diagonal cracks throughout the exterior walls. The

damage was so severe that the building subsequently was demolished.

A seven-story building is seen near the center of figure 4.4.8(b) (location #3). This reinforced concrete

building, which was reported to have built in the 1920's, was severely damaged. The street level view in

'San Francisco Examiner, October 27, 1989

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figure 4.4.11(a) shows that there were many wall openings in the second story. The close-up view in figure

4.4.11(b) shows some of the severe diagonal cracking which formed in the walls between the windows.

The city of Moss Landing, located on Monterey Bay, experienced soil liquefaction at many sites. Serious

structural damage occurred to the State Marine Laboratory. As shown in figure 4.4.12, lateral spreading

of the soil caused extensive cracking of the structure.

In summary, extensive damage was caused by the severe ground shaking in the epicentral region. Houses

fell off of their foundations as a result of cripple stud wall failures. Unreinforced masonry walls cracked

and collapsed. In many cases wall failures lead to roof failures in adjacent structures. Older steel and

reinforced concrete framed structures were damaged. Modern structures and old structures that had been

retrofitted suffered minor or no damage.

4.5 Summary

Damaged buildings were observed within San Francisco, Oakland, and the epicentral region. In San

Francisco, damage was concentrated in the Marina and South of Market Districts. Amplification of ground

motion by deep soil deposits appears to be a primary factor in explaining the distribution of damage. In

the Marina District, the most extensive damage occurred to four-story apartment buildings having soft first

stories and located at street corners. In other Districts of San Francisco and in Oakland, damage

occurred primarily to unreinforced masonry buildings and to older steel-framed buildings with masonry

infills. In the epicentral region, most of the damage occurred to unreinforced masonry buildings and to

wood-framed houses with cripple stud walls. These observations show that, even within a region of the

United States having the highest seismic risk, many seismically inadequate structures exist.

The Loma Prieta earthquake offered opportunities to examine the adequacy of various strengthening

methods that have been used in the San Francisco Bay area. The investigation revealed many examples

of strengthened structures that performed well, but there were also some cases of poor performance. Indepth analyses should be made to determine the performance of repaired structures to obtain basic

information needed to prepare standards for strengthening seismically vulnerable buildings.

Even though the Loma Prieta earthquake was less severe than the "design" earthquake, some modern

structures suffered significant damage. In-depth analyses of the reasons for these failures should be

performed to determine if the failures were due to inadequacies in current standards.

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Figure 4.4.1 Cities visited within the epicentral region.

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Figure 4.4.2 Examples of failures of houses in Los Gatos: (a) large house suffered large distortion and

(b) a small house was rocked off of its foundation.

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y y y ^

\ \ \ \ \ \ \ \ \

t. *>. V. V. \ V V \.

' Cripple Stud Wall Detail

(b)

\ l-i — \ \ \ \ \ \ \ \ \ \ \ *s X \ \ \ \' I* \

\ \ \ \ V. s s s \ \ \

Ground Movement

Schematic of failure caused by inadequately braced cripple stud walls: (a) construction of

cripple stud wall and (b) failure mechanism.

4-46

Figure 4.4.6 St. Patrick's Church in Watsonville was severely damaged.

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Figure 4.4.8 Aerial views of the Pacific Garden Mall: (a) northern portion.

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Figure 4.4.9 Modern reinforced masonry structure which was undamaged; building on the left suffered

roof failure causing two deaths (location #1).

Figure 4.4.10 Historic Cooper House in Santa Cruz was severely damaged and had to be demolished

(location #2).

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Figure 4.4.11 (a) Seven-story, reinforced concrete building suffered severe cracking; (b) Close-up view

of damage to second-story walls.

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5. PERFORMANCE OF BRIDGE AND HIGHWAY STRUCTURES

by William C. Stone, James D. Cooper, and Nicholas J. Carino

5.1 Introduction

The main highway network in the San Francisco Bay region sustained serious damage at several locations.

The most notable is the damage to and collapse of the long, double decked viaduct sections of freeway

in the San Francisco and Oakland areas. Except for the collapse of a single link span of the doubledeck section of the San Francisco-Oakland Bay Bridge, most bridges in the area of the San Francisco Bay

survived the earthquake with relatively minor damage. Most bridges and viaducts had been strengthened

in the California Department of Transportation (Caltrans) Phase I seismic retrofit program which included

identification of structures that are vulnerable to excessive displacements in the longitudinal direction and

have the potential to have spans collapse. Typically, these structures have narrow hinge seats or

discontinuities in the superstructure across piers or abutments and were constructed prior to 1971. Cables

or bars were placed across these joints, tying the elements of the superstructure together.

This chapter describes damage to the Bay Bridge, and to 1-880 (Nimitz Freeway) and other double-deck

elevated highway structures. The results of preliminary analyses to determine the most likely causes of

the collapse of 1-880 are presented below.

5.2 Historical Development of Seismic Criteria

Until relatively recent times, almost all consideration of earthquake forces on structures and relevant code

considerations have been concentrated in building construction as opposed to bridge and highway

construction. The 1906 San Francisco, California earthquake, which caused an estimated $50 million of

property damage, and a subsequent $350 million in fire damage was considered ill fortune (CSSC 1988).

The city was rebuilt in almost identical fashion and remained vulnerable to earthquakes. The 1925 Santa

Barbara, California earthquake caused several million dollars in damage and provided the impetus to

consider adoption of earthquake design provisions in building codes. In 1927, the simple Newtonian

concept of lateral earthquake force being proportional to mass was incorporated in the Uniform Building

Code. The Long Beach, California earthquake of 1933 caused at least $50 million in damage and once

again emphasized the need for more stringent earthquake design requirements. Since then, continual

improvements have been made in the various U.S. codes.

5-1
The first requirement for the inclusion of seismic loading in the design of highway bridges in the United

States was presented in the AASHO, American Association of State Highway Officials (now American

Association of State Highway and Transportation Officials, AASHTO) 1958 and 1959 Interim to the 1957

Specifications (AASHO 1958-1959). These specifications remained unchanged until the 1975 Interim

Specifications. In 1971 the AASHTO Bridge Committee prepared a proposal for a new earthquake

criteria. Prior to adoption, the San Fernando, California earthquake (February, 1971) occurred and

demonstrated that the proposed revisions were inadequate.

The 12th edition (1977) of the AASHTO "Standard Specifications for Highway Bridges" (AASHTO 1977)

presented a new approach for designing highway bridges to withstand earthquake forces. Article 1.2.20 of

the Specification requires, "In regions where earthquakes may be anticipated, structures shall be designed

to resist earthquake motions by considering the relationship of the site to active faults, the seismic

response of the soils at the site, and the dynamic response characteristics of the total structure." The

specification is generally based on the 1973 Earthquake Design Criteria for Bridges by the State of

California. The California criteria were developed for California conditions and subsequently modified to

allow for their use in other areas of the United States where possibly damaging earthquakes can occur.

The impact of the application of the specification, particularly in regions outside California, on design

complexity, design and construction costs, and construction complexity was unknown at the time of

adoption.

Consequently, the Federal Highway Administration initiated a study entitled "Bridge Seismic Design

Guidelines" in 1977 to: (1) evaluate then current criteria used for seismic design; (2) review recent seismic

research findings for potential use in a new specification; (3) develop new and improved seismic design

guidelines; and (4) evaluate the impact of the guides on construction and cost. The guidelines were

completed in 1979 and adopted by AASHTO as a Guide Specification in 1983.

5.3 Current Seismic Design Criteria

A basic premise in developing the current AASHTO bridge seismic design guide specification was that the

provisions should be applicable to all parts of the country. Therefore, for purposes of design, guidelines

were developed for four seismic performance categories (SPC) to which bridges are assigned based on the

seismicity of the area in which the site is located and the importance of the bridge. Bridges are classified

according to their relative importance - either as essential or nonessential. Essential bridges are

determined based on their social/survival and security/defense classification. Essential bridges are those that

must keep functioning during and after an earthquake.

5-2
The Seismic Performance Category defines the differing degrees of complexity and sophistication of seismic

analysis and design which are specified. Four categories are defined. The highest level includes those

bridges designed for the highest level of seismic performance with particular attention to methods of

analysis, design, and quality assurance. The next level includes those bridges for which a slightly lower

level of seismic performance is required, and allows slightly greater potential for damage than for the

highest level. The next level requires a minimum of analysis and draws specific attention to support design

details. Bridges in the lowest category require no seismic analysis but attention is provided to certain

design details for superstructure support. The majority of bridges in the country fall into the second lowest

seismic performance category. Most bridges in California fall into the two highest seismic performance

categories.

The primary basis for development of the seismic design guidelines for bridges is to minimize the hazard

to life and provide the capability for bridges to survive during and after an earthquake with essential

bridges to remain functional. To meet this philosophy, certain principles were followed:

Small-to-moderate earthquakes should be resisted within the elastic range of the structural

components without damage. Realistic seismic ground motion intensities should be used

in the design procedure. Exposure to shaking from large earthquakes should not cause

collapse of all or part of the bridges. And where possible, damage that does occur should

be readily detectable and accessible for evaluation and repair.

In assessing bridge failures of past earthquakes in Alaska, California, and Japan, many of the "loss of span"

type failures are attributed in part to relative displacement effects. Relative displacements arise from out

of phase motion of different parts of a bridge, from lateral displacement and/or rotation of the foundations

and differential displacements at abutments. Therefore in developing the guidelines, the design

displacements were considered to be just as important as design forces. This represented a significant

change in bridge design philosophy. For higher performance category bridges, requirements for ties

between noncontinuous segments of a bridge are specified in addition to minimum bearing support lengths

at abutments, columns and hinge seats.

The methodology used in the AASHTO Guide Specification is in part a "force design" approach but also

addresses the relative displacement problem. The design methodology varies in complexity as the SPC

increases from lowest to highest. Three additional concepts are included in the Guide Specification.

First, minimum requirements are specified for bearing support lengths of girders at abutments, columns,

and hinge seats to account for some of the important relative displacement effects that cannot be

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calculated by current state-of-the-art methods. Second, member design forces are calculated to account

for the directional uncertainty of earthquake motions and the simultaneous occurrence of earthquake forces

in two perpendicular horizontal directions. Third, design requirements and forces for foundations are

intended to minimize damage since most damage that might occur will not be readily detectable. The

selection of ground motion intensities to be used with the seismic design provisions was carefully reviewed

by the Applied Technology Council (ATC 1978). Considerable study and effort had recently been made

to develop seismic risk maps and associated design spectra for the "Tentative Provisions for the

Development of Seismic Regulations for Buildings (ATC-3-06)." The maps are based on an appraisal of

expected ground motion intensities, the probability that the design ground shaking will be exceeded is

approximately the same in all parts of the United States, and frequency of occurrence of earthquakes in

various regions of the country. It is possible that the design earthquake ground shaking might be

exceeded, although the probability of this happening is quite small, about 10 percent in 50 years.

The AASHTO Guide Specifications provide for two methods of analysis which vary according to the

refinement in the mathematical idealization. They are the single-mode spectral analysis method and the

multi-mode spectral analysis method. All methods assume simultaneous support excitation.

5.4 Approach for Improved Seismic Response

The highway network is a vast system which forms vital and essential links between cities and towns across

the United States. The interstate system of highways extends approximately 41,000 miles (66,000 km) and

has about 47,000 bridges in the network. Added to that is another 89,000 bridges on the primary system.

The exposure of the network of roads and bridges to seismic hazards varies greatly across the country.

Bridges, grade separation, and elevated highway structure form critical links in this network and are most

susceptible to seismically induced damage. They also represent the greatest economic risk if destroyed or

damaged.

Two approaches are being taken to improve the seismic resistance of highway bridges. The first is to

design and construct new highway bridges for seismic resistance. This requires considerable time, but is

economically reasonable, and is being pursued. Design guidelines are being upgraded as more knowledge

is gained about the response of specialized transportation structures to seismic activity. These new design

guidelines can be applied to new construction as older bridges are removed from service because they are

either structurally unsound or functionally obsolete. The second approach involves identifying those

existing bridges which are important to the network and are susceptible to significant damage or collapse

in the event of an earthquake. Those structures can be strengthened to enhance their response to seismic

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activity. This approach, however, requires significant expenditures and can prove economically difficult.

A balanced approach is needed to reduce earthquake hazards. Design and construction of new structures

should use available, cost effective design standards.

5.5 Introduction to Viaducts and Bridge Performance

Following the Loma Prieta earthquake Caltrans engineers conducted preliminary inspections of more than

1500 bridge structures in the area affected by the earthquake and determined that some 73 bridges had

suffered minor damage of varying degrees, that five major viaducts and five other bridges suffered

significant structural damage, and that major or partial collapses occurred at three sites. Figure 5.5.1

depicts the locations of the five damaged viaducts and the collapsed span of the Bay Bridge, indicated

by shaded circles, and the 1-880 collapse by the shaded oval. From previous discussion of the

geotechnical aspects of the earthquake in Chapter 3, it can be appreciated readily that there is a strong

correlation between the location of damage and the presence of deep underlying deposits of natural or

man-made overburden. The three major collapse sites at 1-880, the Bay Bridge, and Struve Slough, are

of particular interest and will be discussed first. Of these, the first two were visible symbols of the Loma

Prieta earthquake with 1-880, in particular, capturing the attention of the national media for days

following the quake as efforts to free trapped motorists continued. Other viaduct failures included the

Southern Freeway [1-280] just to the east of Highway 101; 1-280 at the China Basin distribution structure;

The Central Freeway [Highway 101] in downtown San Francisco; the Embarcadero Freeway [1-480] at

Mission Street in San Francisco; and 1-980 just to the east of 1-880 in Oakland. None of these viaducts

collapsed. They did, however, suffer sufficient damage that all are still closed and have required the

installation of significant timber shoring to insure stability until a proper retrofit can be effected.

Damage to structures at these sites is described in sections 5.9 through 5.11.

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5.6 The 1-880 Collapse

Interstate 880, also known as the Nimitz Freeway, is a primary north-south oriented 8-lane highway

connecting San Jose with Interstates 580 and 80, adjacent to the east end of the San Francisco-Oakland

Bay Bridge. Generally, 1-880 is a ground level freeway with the exception of a 2 mile (3 km) elevated

segment which is oriented approximately north-south and is bounded by 7th Street on the south and 34th

Street on the north in Oakland (see fig. 5.6.1). The configuration of the elevated portion, known as the

Cypress Structure, consisted of a bi-level system with four lanes of north-bound traffic on the first level,

approximately 25 feet (8 m) above ground level, and four lanes of south-bound traffic on the upper level,

approximately 50 feet (15 m) above ground level. Surface level traffic paralleled the Cypress Structure on

both sides with underpasses at the major cross streets. Design work for this section was begun in 1951

and construction was completed in 1957. It is to be noted that the AASHTO design specifications for

highway structures for this period (AASHTO, 1953) make no reference to design requirements for resisting

lateral loads. However, the Caltrans Bridge Department design supplement for earthquake loads (Caltrans,

1949) did require bridge structures to resist earthquake loads of 0.06 times the weight of the structure for

bridges on pile foundations as used for the elevated portion of 1-880.

The structural configuration of the Cypress Structure consists of a series of 124 reinforced concrete

transverse bents which support longitudinal cellular box girders that carry the road deck. The box girders

vary from 71 to 90 feet (22 to 27 m) in length by 55 feet (17 m) wide and are integrally cast flush with

the tops of the bent girders as shown in figure 5.6.2. The road deck has a relatively uniform weight of

approximately 8 tons per foot (230 kN/m) along the longitudinal direction, which can be used to determine

vertical and inertially induced loads to the various bents.

5.6.1 Description of Structure

There were 11 distinctively different bent configurations used on the Cypress Structure, but only three

primary types were present in the majority of the collapsed section, which extended from the first

expansion joint north of Bent 63 at 18th Street to just short of Bent 113 at 34th Street. The total length

of the collapsed section was 3970 feet (1210 m), contrary to the 1.5 mile (2.4 km) figure commonly

quoted in the media following the quake. The three predominant bent configurations are shown in figures

5.6.2 through 5.6.4 and are labelled as Type 1, 2, and 3 bents. A fourth type, which was similar to the

Type 3 bent but with an additional support column for the first level and prestressing for the upper bent

girder, is shown in figure 5.6.5. A fifth variant, similar to a Type 2 bent, but with a

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Table 5.6.1 Bent Configurations for the Collapsed Portion of 1-880

Bent Type Bent Type Bent Type Bent Type Bent Type

63 1 73 5 83 1 93 1 103 1

64 1 74 5 84 1 94 1 104 1

D J

1X / J 9 o ^1X QS A ins 1

j_ 66 1 76 2 86 1 96 4 106 1

67 1 77 2 87 1 97 4 107 1

68 1 78 2 88 1 98 4 108 1

69 1 79 2 89 1 99 1 109 6t

70 3 80 2 90 1 100 1 110 6t

71 2 81 1 91 1 101 1 111 6t

11 2 82 1 92 1 102 1 112 7$

f Bottom portal frame detail is similar to Type 4 bent; Top portal frame details are similar to Type

1 bents. The roadway deck for the lower bent gradually decreases in height above the surface from

Bent 109 to Bent 111.

* This consists of a single portal frame supporting the upper roadway with hinged connections between

the column bases and the pile caps and a moment resisting connection between the beams and the

transverse girder. The lower roadway is supported a few feet above the surface by another transverse

girder with moment resisting connections to the columns.

cantilevered detail for the lower level girder, is shown in figure 5.6.6. A summary of bent types within

the collapsed section is presented in table 5.1.

There are several characteristics common to most bents for the Cypress Structure. In all cases the lower

portion of each bent, comprising two vertical support columns and a horizontal (some were superelevated)

connecting girder, was designed to achieve a moment resisting connection between the columns and girder.

The columns generally measured 72 inches (1.83 m) deep by 48 inches (1.22 m) wide and were heavily

reinforced with 44-#18 Grade 40 longitudinal reinforcing bars. Contrasting this massive axial

reinforcement, #4 bar rectangular ties with 90-degree hooked ends were provided on 12 inch (310 mm)

centers as lateral confinement. The columns were supported by 36 to 54 inch (0.91 to 1.52 m) thick

reinforced pile caps which in turn were cast on top of driven pile foundations. The pile tip penetration

depths varied from 16 feet (5 m) at Bent 63 (the southernmost limit of the collapse) to a maximum of

55 feet (17 m) for Bents 72 through 78. The majority of those from Bent 88 through the northern limit

of the collapse had depths of 40 to 45 feet (12 to 14 m). The connection between the columns and the

pile caps was designed as a rectangular depressed shear key, generally measuring 18 inches (460 mm) long

by 36 inches (910 mm) wide and containing 4-#10 dowel bars which formed a 10 inch (250 mm) square

at the center of the key. The remainder of the joint was filled with a half-inch (13 mm) layer of

expansion joint material. For all intents and purposes this detail acted as a hinge. Several bents, notably

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95-98 which are skewed on a 29-degree angle at Grand Avenue to accommodate railroad tracks, have

longer spans which dictated the use of a third central column as shown in figure 5.6.6. The lower cross

girders generally had dimensions of 48 inches (1.22 m) in width with a depth varying from 96 to 114

inches (2.44 to 2.90 m). Typical reinforcement details are shown in figure 5.6.6, Section C-C. The details

of the upper bents are more varied. Type 1 bents, which comprised more than half of those in the

collapsed section, were designed similarly to the lower bents, that is, moment resisting connections between

the columns and cross girder, and pinned connections at the base of both columns. The upper columns

were tapered from 48 inches in width at the top, to accommodate girder reinforcement, to 36 inches (910

mm) at the bottom with a constant 48 inch section depth. Like the bottom columns, these were heavily

reinforced in the axial direction with 22-#18 bars, half that of the lower column, with #4 ties with 90-

degree hooks on 12 inch (310 mm) centers. Of particular interest is the detail of the connection between

these upper columns and the lower bent frame shown in figure 5.6.7. A 27 inch (680 mm) high pedestal,

reinforced as shown in Section B-B' in figure 5.6.7, was used to accommodate placement of the shear

key/hinge detail shown in Section A-A'. The 4-#10 dowel bars shown in the shear key detail extended

30 inches (760 mm) above and below the joint which, under normal circumstances, meets the development

length requirements for this type of bar. It should be noted that the #4 stirrups used for lateral

confinement of the lower bent columns, observed during the field investigation, stop at the bottom face

of the lower bent girder, i.e. there was no specific confinement steel within the joint. The 14-#10 x 36-

inch (910 mm) pedestal reinforcement bars shown in Section B-B' were embedded just 6 inches in the

upper, outside corner of the lower bent beam-column joint, in the same area where some 16-#18 negative

moment reinforcement bars for the lower bent girder splice to two rows of vertical #18 bars for the lower

column. The congestion caused by this detail is evident just below Section B-B' in figure 5.6.7. Lacking

substantial lateral confining reinforcement in the joint, it appears likely that any lateral loads transmitted

across the pedestal (Section B-B') would be resisted only by the shear strength of the concrete and that

negligible resistance to lateral motion would be contributed by the reinforcement contained within the

pedestal. Calculations, presented below, of the loads to cause failure, which assume full development of

the pedestal ties, can therefore be considered conservative.

The design of the Cypress Structure permitted a future access ramp for Grand Avenue to be added to the

west side of the freeway. In order to accommodate the extended spans required of the upper bent girder,

prestressing tendons were used in place of deformed steel reinforcement. These prestressing bars, which

were unbonded, were designed to be extended and subsequently anchored to the west face of the girder

following the addition of the ramp. The three-hinge detail associated with these prestressed bents is

shown in figure 5.6.3. The two upper hinges were similar in detail to that shown in Section A-A' of

figure 5.6.7. The presumed intent of this approach was to accommodate shortening of the upper deck

5-9
girder, apparently assumed at the time to be on the order of 2 to 3 inches (50 to 80 mm). Recent

calculations by Caltrans engineers indicated a shortening of 5/8 inch (16 mm), but this is based on nearly

40 years of experience. At the time of design, the concept of prestressing was in its infancy and the use

of a dual hinged column on one side of the bent to accommodate the presumed large displacements

induced by prestressing is an understandable approach where gravity loads are the predominant force acting

on the structure. With the three-hinge design, however, all lateral loads must be resisted by the righthand (east side) column alone (see figs. 5.6.3 and 5.6.6) which is designed with continuous longitudinal

reinforcement that extends upward from the lower bent column. The notable exception to this situation

is for Type 5 bents where a cantilevered detail demands that the upper column steel be anchored only in

the lower bent girder.

The specified working stress for the concrete was 1,250 psi (8.62 MPa). Cores taken after the earthquake

from column sections by researchers from the University of California at Berkeley (EERC, 1989) indicated

a compressive strength of 6,0{X) psi (41 MPa) and a modulus of 3.7 million psi (25.5 GPa). The

reinforcement, except for prestressing bars, was Grade 40.

In 1977, as part of the Caltrans Phase I Seismic Retrofit Program, the Cypress Structure was equipped

with cable restrainers at each transverse expansion joint between the longitudinal box girders. These

expansion joints were spaced at approximately 240 feet (73 m) and were intended to reduce temperature

induced stresses in the deck structure. There were, on the average, three bents between expansion joints

which could be expected to act as a continuous unit along with the upper- and lower-deck box girders.

In order to prevent earthquake-induced ground displacements from unseating one of these expansion joints,

which would lead to immediate collapse of at least one span, a series of cables were used to anchor the

ends of the mating box girders together. Figure 5.6.8 shows a typical detail in which three cable restrainer

units were installed at each expansion joint. In order to prevent puUout of the cable assemblies under

seismic loads, a set of reinforced concrete bolsters were cast to increase the thickness of the end

diaphragms for the box girders. Seven 3/4 inch (19 mm) wire rope cables were wrapped around a metal

drum on one side to form a horseshoe-shaped cable bundle. The ends of the cables were swaged into two

7-cable anchor blocks located on one side of the expansion joint. It may be appreciated that the

presence of these restrainers, in addition to preventing the unwanted dropping of a deck segment, also

effectively creates a longitudinal hinge at each expansion joint, across which both vertical and horizontal

structural loads can be transmitted.

5-10
5.6.2 Description of Collapse by Motorists

It was fortunate that the third game of the World Series of baseball, between the San Francisco Giants

and the Oakland A's, was about to be played in San Francisco at the time of the earthquake. It has

been reported by California state transportation officials that traffic on 1-880 was extremely light as a result

of many people having left work early to watch the game on television or at the stadium. Furthermore,

due to road construction south of the Cypress Structure, a substantial amount of north-bound traffic was

detoured through alternate ground-level roads^ When the earthquake occurred, a 3970-foot (1210-m)

section of the double-decked Cypress Structure collapsed, killing 42 motorists, most of whom were

travelling north-bound on the lower level and were crushed by the upper deck. Useful accounts of the

collapse, by eyewitnesses who were in a position to ascertain the dynamic behavior of the structure, are

scarce. One pedestrian near 18th Street observed, "It was like a big, giant, long ocean wave, and behind

each wave a portion of the freeway collapsed. As the top level crashed down on the cars below, shearing

cleanly at 18th street, cars shot off the sides of the [upper] freeway, crashing and flipping into the street."^ One survivor who was on the upper deck travelling south said that, "there was a distinct initial [sideways]

jolt as the earthquake first struck. I thought one of my tires had gone flat. My first thought was to get

off the freeway as quickly as possible, so I sped up. Approximately 20 seconds later the second wave hit

and behind me [in the mirror] I could see concrete dust plumes shooting into the air."^ This particular

witness was one of the last to make it past 18th Street ahead of the collapse. The same witness also

made the poignant remark that in retrospect he noticed that others were making all efforts to stop their

cars and that lane changes were being made involuntarily. Furthermore, it was suggested that many of

those travelling north-bound intentionally stopped beneath girder bents, under the false assumption that

these, like the doorways of buildings which purportedly offer greater safety in the event of an earthquake,

would provide safety from falling objects. Several survivors indicated that sufficient time was available

prior to the collapse that motorists were able to stop their cars and, in some cases, even get out.

Unfortunately, and as the clustering of car locations outlined in paint on the upper deck by recovery

personnel attests, parking beneath the bents was likely the worst choice, as there was no clearance when

the girders fell.

^Governor's Board of Inquiry Meeting, State of California: "The 1989 Loma Pricta Earthquake,"

December 13-14, 1989, Hyatt Regency International Hotel, Oakland, CA

^San Francisco Examiner, October 18, 1989 Special Edition.

'Governor's Board of Inquiry Meeting, State of California: "The 1989 Loma Pricta Earthquake,"

December 13-14, 1989, Hyatt Regency International Hotel, Oakland, CA

5-11
5.6.3 Summary of Aerial Photos

The second day after the earthquake an aerial reconnaissance of the 1-880 collapse was carried out via

helicopter by the team. A set of close-up oblique aerial photos of the collapsed portion of the Cypress

Structure is presented in figures 5.6.9 through 5.6.22. Figure 5.6.9 shows an overall view looking south

fi-om Bent 113. The northern extent of the collapse is visible in figure 5.6.10 with Bent 113 remaining

intact at left center. Beginning at Bent 106 (fig. 5.6.11) and extending through Bent 104 (fig. 5.6.12),

both upper and lower levels of the freeway collapsed to ground level. From Bent 104 through Bent 98,

only the upper deck collapsed. Figures 5.6.13 through 5.6.15 show the anomalous span between Bents 96

and 97 which remained intact following the earthquake. In viewing these figures, it should be kept in

mind that Bents 95 through 98 were all Type 4 bents and that Bents 96 and 97, specifically, were skewed

to the primary axis of the bridge by 29 degrees to accommodate the railroad tracks visible in figure 5.6.14;

Bents 95 and 98 were not skewed. Furthermore, expansion joints exist immediately north of Bents 95 and

98.

Various collapse sequences have been proposed centering on span 96-97. One theory assumes that span

96-97 served as the initiation point, with two progressive collapse waves propagating to the north to Bent

113 and simultaneously to the south to Bent 62. Another scenario involves span 96-97 serving as the

arrester, with simultaneous collapse waves propagating towards it fi-om Bent 113 on the north and from

a point somewhere in the prestressed section between Bents 71 and 80 to the south. The southern

collapse section would, consequently, have also propagated south to Bent 62. Yet another theory holds

that the lateral accelerations experienced by Bents 96 and 97 were insufficient to cause the shearing

failures seen at Type 1 and 2 bents and that all of these latter bents essentially failed nearly

simultaneously. Unfortunately, available eyewitness descriptions of the collapse are not precise enough

to validate these theories, or to suggest others. However, the observation by several survivors that the

entire structure was moving up and down as though "giant waves" were travelling down the structure, and

the fact that the upper deck came to rest squarely on top of the lower deck (implying nearly simultaneous

failure of east and west upper bent columns) are of particular interest.

Figures 5.6.16 through 5.6.18 show a uniform section between Bents 94 and 80 in which the upper deck

has collapsed directly on top of the lower deck. All of the bents in this section were Type 1 and

exhibited a distinctive failure mode in which both upper bent columns sheared at their base from the

supporting pedestals and splayed outward as the weight of the upper deck carried the bent down to the

lower level girder where the collapse was generally arrested. Beginning at Bent 80 and extending south

to Bent 71 (figs. 5.6.18 and 5.6.19) all bents were of the three-hinged post-tensioned variety (Types 2 and

5-12
5), designed to be extended westward for the future construction of the Grand Avenue access ramps.

These three-hinge bents exhibited two distinctively different collapse modes. The first, visible between

Bents 74 and 80 in figures 5.6.18 and 5.6.19, involved the destruction of the eastern upper bent columns

and the general rotation of the upper deck in a clockwise fashion (looking north), pivoting about the

western two-hinged upper bent columns. In stark contrast, the western two-hinged columns were ejected

from the structure for Bents 71 through 73. These column segments landed as much as 30 feet (9 m)

from the bent face, as evidenced by 2 foot (600 mm) deep, 6 foot (1.8 m) diameter craters in the adjacent

pavement where the column ends impacted. Bents 63 through 69 (figs. 5.6.20 through 5.6.22) were all

Type 1 and exhibited failure similar to that described previously. The southern extent of the collapse can

be clearly seen at the expansion joint just north of Bent 62.

5.6.4 Ground Details

Figure 5.6.23 shows a ground-level view looking northward from Bent 86 on the west side of the

expressway. All of the bents visible in this photo were Type 1 which invariably exhibited a symmetrical

failure with both upper columns shearing free at their base and being splayed outward as the upper deck

fell. The timber posts visible beneath the lower deck were installed by Caltrans under each bent to insure

that there was no subsequent settling of the structure during rescue operations. Further evidence that the

upper deck fell nearly squarely on the lower deck with no significant sidesway is shown in figure 5.6.24

which is looking northward from atop Bent 80.

Figures 5.6.25 through 5.6.28 show four minor variations of Type 1 bent failure. All views are from the

west side of the respective bents. At Bent 65 (fig. 5.6.25), it can be seen that the upper column support

pedestal is still attached to the base of the column and that an inclined failure plane exists through the

failed pedestal. Note that confining reinforcement in the vicinity of the lower girder/column joint is nonexistent. As for most Type 1 bent failures, the upper columns remained tenuously attached to the upper

bent girder. Bent 85 (fig. 5.6.26) shows a similar situation regarding the upper column support pedestal

(beneath the numbers "85" in the figure). Here it is evident that the failure plane for the pedestal ran

flush with the curved face formed by the lower bent girder negative moment steel. It is also evident that

the seismic loads, in combination with a lack of confining steel in the upper beam/column joint, were

sufficient to induce an explosive rupture of concrete at the top of the column, breaking it free from the

upper girder steel seen at the top center of the photo. It should be emphasized, and it will become clear

in subsequent discussion, that such compressive failures did not initiate the collapse of the bent, despite

the lack of lateral confining steel in the upper girder/column joint. At Bent 90, the upper column

apparently sheared through the 4-#10 dowel bars at the center of the shear key and was propelled outward

5-13
with sufficient force to separate the column top from the upper bent girder. The upper bent column

support pedestal can still be seen projecting above the lower column.

Figure 5.6.28 shows a final variation of a Type 1 bent failure. This particular column, at Bent 94, carried

a pipe through its center which drained the upper deck. Given a lack of significant confining

reinforcement (note the thin #4 ties which appear as wires compared with the massive #18 axial

reinforcement), the presence of this pipe likely served to create the fracture path which split the column.

Figure 5.6.29 shows a closeup of a Type 2 bent failure in which the eastern half of a series of threehinged prestressed bents collapsed, due to shearing failure at the base of the east column, leading to a

rigid body rotation of the upper deck about the western, two-hinged columns. As can be seen, the western

column shear keys did perform as hinges even under rotations sufficient to cause compressive spalling on

the interior face of the columns (fig. 5.6.30). Most of the upper bent girders fractured due to impactinduced bending stresses. This in turn yielded, and subsequently fractured most of the post-tensioning bars

which, because they were unbonded, were partially expelled from the girder due to rebound from the

release of stored elastic energy, as sown in figure 5.6.31. It is of some significance that Bent 78, the

innermost frame in figure 5.6.32, was skewed at an angle of 16 degrees. This difference, although subtle,

may be part of the explanation for why this particular section did not fail completely. This performance

is to be directly contrasted with Bent 72 (see figs. 5.6.33 and 5.6.34) in which the western two-hinged

columns were ejected from the structure. Of particular note in figure 5.6.33 are the bent #10 bars

projecting both from the top of the lower bent column as well as the bottom face of the upper bent

girder. These represented the only continuous reinforcement through the upper and lower shear keys.

Figure 5.6.35 leaves no doubt that lateral confining steel was insufficient to prevent localized fracture of

the concrete and the subsequent pullout of these dowels. A closeup of the top of the Bent 71 upper twohinge column in figure 5.6.35 shows a detail of the still intact shear key, 4-#10 dowel bars, and, in this

particular instance, a central drainage pipe. In this case the dowel bars remained intact within the column

and pulled out, instead, from the upper girder joint.

Figures 5.6.36 and 5.6.37 show details of the span between Bents 96 and 97, the only one to remain

completely intact between Bents 62 and 113. There are several obvious differences which could account

for this. The Type 4 bent, which includes Bents 95-98, contains an additional central column in the lower

bent which would serve to limit lateral sway at the first deck level. More importantly, the reinforcing steel

at the base of the upper bent columns is continuous with the lower columns, eliminating the pedestal/hinge

detail that appears to have contributed to widespread failure in Type 1 bents. Figure 5.6.37 indicates

that, despite severe flexurally induced cracking at the lower column/girder joint, the upper and lower

5-14
columns remained intact. These structural differences contributed to increase structural stiffness and thus

decrease the natural period of vibration. Determination of whether the differences in vibrational

characteristics between bents is of vital significance in understanding the overall collapse is beyond the

scope of this report. However, it would appear that, due to the long length of the structure with respect

to the propagation speed of the disturbance (it has been estimated that a lag time of several seconds

existed between the time when the surface waves hit the southern end of the Cypress Structure and when

they arrived at its northern end), such variations in structural vibrational characteristics, as well as those

for the underlying soil at each particular bent, may need to be considered for an accurate dynamic model

of the structure.

The section of expressway between Bents 104 and 106 (fig. 5.6.38) contains the only portion of the Cypress

Structure to have exhibited complete failure in both upper and lower decks. These were Type 1 bents

in an area where the lower road deck was beginning its descent to the northern distribution structure.

Beyond Bent 106 both upper and lower decks were superelevated on the east side. Other than these

minor differences, there were no particularly distinguishing characteristics which would have made these

bents more susceptible to failure. It has been suggested (EERI, 1989) that greater accelerations may have

been experienced towards the northern end of the structure due to the presence of Bay Mud deposits,

which are deepest towards the northern end of the collapsed section and would have led to soil

amplification. This argument is unconvincing in light of geotechnical data subsequently made available

(CDMG, 1989b). The remaining possibilities include construction defects and/or resonance of the entire

Cypress Structure which could have led to higher forces in this particular area, as a result of random

coincidence of modes, than would be expected using the simple Newtonian lateral force approach

discussed below. In any event, there was vastly different performance of the lower bent columns to the

north of Bent 104, as exhibited, for example, at Bents 105 and 108 shown in figures 5.6.39 and 5.6.40.

Bent 108, in particular, exhibited the only gross shear failure of a lower column seen in the entire Cypress

Structure.

Figure 5.6.41 shows a view looking south from Bent 114 at the northern extent of the collapse. Significant

changes in bent configuration begin to take place at Bent 113. The lower deck is supported nearly at

ground level, while the upper deck is supported on outrigger beams cast integrally with their supporting

columns. No hinges, except at the foundation level, are present in these bents. This change in stiffness

at the commencement of the northern distribution structure was certainly a factor as to why the collapse

either initiated or terminated at this location. Similar changes in structural configuration exist at the

southern limit of the collapse where the presence of access ramps led to stiffer bents.

5-15
5.6.5 Observations on Contributing Causes of Collapse

There is no site-specific acceleration record for the Cypress Structure. However, initial reports from the

California Department of Conservation, Division of Mines and Geology (CDMG, 1989a), and from the

U.S. Geological Survey provide strong motion records from the Oakland Outer Harbor Wharf (1.5 miles

(2.4 km) to the west); at the base of a two-story office building in Oakland (1.5 miles to the southeast);

and from the base of another two-story building in Emeryville (1 mile (1.6 km) to the north).

Acceleration records from these three sites had peak horizontal accelerations of 0.29g, 0.26g, and 0.26g,

respectively. In a second report on the geotechnical aspects of the earthquake by the California

Department of Conservation, Division of Mines and Geology (CDMG, 1989-b) it is observed that:

"The strong motion data from the Oakland area does indicate that amplified levels of motion occurred

at the Cypress Structure. However, the amplification effect occurs throughout the flat lying portions

of Oakland, not just in a localized zone near the Cypress Structure.... Whatever factor is causing the

increased shaking levels in Oakland extends throughout the flat lying areas, and is not limited to areas

of bay mud."

On the basis of these findings it may be concluded that the Cypress Structure was subjected to horizontal

accelerations of between 0.26g to 0.29g. Acceleration records from these sites taken in perpendicular

directions do not differ greatly, indicating that the orientation of the Cypress Structure was not of critical

importance in its subsequent response to the earthquake. Response spectra for the Oakland Wharf site

(CDMG, 1989-b) indicate peak structural accelerations of Ig for a structure with a natural period of

vibration of 1.5 seconds at 2% damping. In order to estimate the natural period of the Cypress Structure,

a series of finite element computer models were constructed for Type 1 and Type 2 bents. The models

included a single bent of the specified type which included the inertial mass and stiffness contributions of

an 80 foot (24 m) section of the deck box girders as well as those for the bent girders and columns.

The results of these analyses, based upon uncracked section properties derived from the as-built drawings,

are presented in figures 5.6.42 and 5.6.43. The results indicate first and second mode frequencies for Type

1 bents of 2.5 Hz and 6.4 Hz, respectively; for Type 2 bents the first and second mode frequencies were

2.0 Hz and 6.3 Hz, respectively. The values for the Type 1 bent compare extremely favorably with first

and second mode frequencies of 2.5 Hz and 6.5 Hz, respectively, measured by University of California at

Berkeley researchers during a full scale test of Bents 45, 46, and 47 following the earthquake". On the

''Governor's Board of Inquiry Meeting, State of California: "The 1989 Lonia Prieta Earthquake,"

December 13-14, 1989, Hyatt Regency International Hotel, Oakland, CA.

5-16
basis of these analyses, and the response spectra for the Oakland Outer Harbor Wharf station, it can be

determined that for 2% damping (as determined from the Berkeley tests) both Type 1 and Type 2 bents

in the Cypress Structure experienced first mode equivalent accelerations of 0.35g.

Following the Uniform Building Code (UBC, 1988) assumption of an inverted triangular distribution for

the effective lateral load due to the earthquake applied to each deck level, and taking into account that

the weight of each level is approximately equal, and that the upper roadway is approximately twice the

height above the ground as the lower level, the elastic force at the upper level will be twice that at the

lower level. Thus two-thirds of the total base shear force is effectively applied to the upper level bent and

one-third to the lower bent. These forces were subsequently used as input for static finite element

analyses of bent Types 1, 2, and 5, in which the three predominant types of failure were observed.

Vertical inertial loads were applied as a uniform acceleration of 1.06g, based on the vertical peak

acceleration components at the seismic stations previously mentioned.

For lightly reinforced concrete sections, plots of linear elastic tensile stress trajectories are good indicators

of where cracks will initiate and the likely paths they will follow under a decreasing-strength failure

situation. The results, presented in figures 5.6.44 through 5.6.46, both as contours of maximum principal

tensile stress and as tensor plots of the principal stresses, indicate that the column support pedestal for

the upper columns in Type 1 bents is highly stressed and that crack initiation will occur beneath the shear

key with the crack pattern as shown in figure 5.6.44(d). Crack initiation in the concrete was assumed to

occur at a tensile stress of 460 psi (3.17 MPa), based upon the ACT tensile strength formula (ACT, 1989)

for a compressive strength of 6,000 psi (42 MPa). The stresses shown in figures 5.6.44 through 5.6.46 are

a result of both the lateral seismic load as well as outward shearing forces associated with the portal frame

reactions at the shear key due to vertical loads on the upper bent. This latter contribution accounts for

more than half the load required to shear the column support pedestal for a Type 1 bent, which is

determined to be 333 kips (1480 kN) for concrete having = 6000 psi and is directly dependent upon

the maximum vertical acceleration seen by the structure. The calculated lateral base acceleration required

to achieve a shearing load of 333 kips at the pedestal, by modifying Priestley et al. (Priestley, 1989) to

account for the measured concrete strength, is 0.2g. With the cracking of the right column support

pedestal, it can be seen that the upper left-hand beam/column joint is subjected to large tensile stresses

due to bending which would lead to crack initiation. While there is considerable positive moment steel

in the upper bent girder at this location, no transverse reinforcement was noted during the field inspection.

This condition could be expected to contribute to a brittle, explosive failure mechanism following shearing

of the left lower pedestal as the bent was rocked in the opposite direction on the reverse cycle. Based

on this reasoning, it has been hypothesized by some investigators that the Type 1 upper bents could have

5-17
been destroyed during just one complete cycle of loading at maximum acceleration (Priestley, 1989; EERI,

1>89).

Static load tests carried out on Bents 45-47 by the University of California at Berkeley^ indicated failure

initiation in Type 1 bents to take place at a lateral load of 465 kips (2070 kN) per bent, with the lateral

load being applied to the top deck of the upper bent. Utilizing the UBC load distribution approach

previously described, this leads to an equivalent base shear of approximately 700 kips (3120 kN) which

equates, for a typical bent and 80 feet (24 m) of deck, to a base acceleration of 0.25g.

An equally serious situation exists for Type 2 bents, as shown in figure 5.6.45. Cracking is expected at

the joint between the upper and lower columns along the diagonal path shown. Likewise, flexural cracking

is indicated at the interior corner of the lower bent beam/column joint. Due to the increased cross section

of the upper columns, the collapse load is expected to be higher than for the Type 1 bents. The

calculated collapse load, following Priestley et al. (Priestley, 1989) and modified for a concrete compressive

strength of 6000 psi, indicates failure at a shear force of 451 kips (2010 kN), or 38% higher than that

for a Type 1 bent. The calculated collapse acceleration for Type 2 columns is 0.25g; no corresponding

experimental data are available.

Figure 5.6.46 shows the stress distribution at the critical joint between the upper bent column and the

cantilevered portion of the lower bent girder for Bent 74. Although it is indicated that a vertical crack

will initiate under the given loads at the interior corner of this joint, this would be insufficient to cause

failure given the massive amount of negative moment steel from the lower bent girder at this location.

Field inspection described in the report by Bertero et al. (Bertero, 1989) indicates that the negative

moment steel did not extend sufficiently far into the joint to be effective. It is hypothesized that a shear

failure occurred at this location immediately following the formation of the tensile crack shown at the

interior corner. The formation of this crack, based upon a linear scaling of the applied lateral loads,

could initiate at a lateral acceleration as small as O.lg.

The above discussion is graphically illustrated in figures 5.6.47 through 5.6.49 which indicate the proposed

failure initiation sequence for the three predominant types of bent failures observed at the Cypress

Structure and depict the state observed in the field after the collapse.

'Governor's Board of Inquiry Meeting, State of California: "The 1989 Loma Prieta Earthquake,"

December 13-14, 1989, Hyatt Regency International Hotel, Oakland, CA

5-18
It is interesting to note that the calculated column shear resistance reported by Priestley et al. (Priestley,

1989) for Type 4 bents (see fig. 5.6.5), when modified to account for the measured compressive strength

of the concrete, comes to 398 kips (1770 kN). Accounting for portal frame reaction shear at the column

bases for the upper bent, the lateral base acceleration required to achieve failure is 0.38g, i.e., greater than

the expected acceleration from the response spectra reported by CDMG (CDMG, 1989b). These frames,

as can be seen in figure 5.6.37, suffered significant damage, but remained standing.

The analyses described thus far assumed that each bent and its tributary sections of the box girder deck

respond to the earthquake in a manner which is independent and decoupled from the transient response

of the adjacent bents. The analyses also assumed a failure mechanism triggered by the effects of lateral

loading. Two pieces of evidence are worth considering. First, the failure was essentially vertical. It

is unlikely that such uniform vertical collapse would have resulted solely from the effects of lateral loading.

Second, eyewitness reports indicated the presence of "giant waves" rolling down the length of the elevated

structure. Taking these two points into account, dynamic modal analyses of a 9-span segment of the

Cypress Structure were carried out. The results indicate that the 9-span long structure had a lateral

vibrational period of 9 seconds, well beyond the period of peak lateral accelerations based upon the

response spectra of the record from the Oakland Outer Harbor Wharf site. However, the vertical mode

of vibration (see fig. 5.6.50) had a period of about 0.16 seconds (6.2 Hz) which is remarkably close to the

predominant period of the vertical acceleration of the record for the area. Structural amplification in this

mode could account for three disparate aspects of the 1-880 collapse: 1) it accounts for the observations

made by witnesses; 2) structural amplification in this mode would lead to large vertical forces acting upon

the bents. It is important to recall that more than half the shearing load to cause failure in a Type 1

bent was due to portal frame horizontal reactions which are a result of vertical loading. A significant

increase in the vertical reaction forces, as a result of resonance in the mode shown in figure 5.6.50, would

also lead to failure of the bents, but with the important exception that the failure would necessarily be

symmetric; 3) the lower and upper road decks do not necessarily vibrate in phase, nor with the same

amplitude (see fig. 5.6.50). For those places where the decks vibrate in phase, the reaction forces could

be much greater. This helps to explain the variations in failure which were observed along the length of

the collapsed section, particularly towards the northern end where both decks collapsed.

The above analyses, which are of a preliminary nature, explain the possible causes for the 1-880 collapse.

It is concluded that simple lateral load analysis would not be adequate to describe completely the seismic

response of the Cypress Structure and that the dynamic behavior of the entire structural system needs to

be considered in order to predict correctly the modes of failure.

5-19
5.6.6 Summary

1. While the precise dynamic behavior of the Cypress Structure prior to its collapse remains to be

explained, it may be concluded that at least three different failure mechanisms existed whose ultimate

resistance could be expected to be substantially exceeded during the earthquake. Type 1 bents were

shown experimentally to fail at 0.25g lateral acceleration, significantly lower than the anticipated 0.35g

acceleration predicted for the site based on available response spectra. The inability to resist the

applied lateral load for all bents which collapsed can be directly traced to a lack of sufficient

confining steel within the critical joint details. It should, however, be appreciated that the calculated,

and measured, resistances of the bents to lateral load substantially exceeded the AASHO design

criteria under which the bridge was designed in 1951.

2. The Cypress Structure was non-redundant. Once one of the upper bent columns for a particular bent

failed, the full load would be transferred to the remaining support, leading to an explosive failure

consistent with that observed at the site following the earthquake.

3. The lower bent columns and girders appeared, in general, not to be a factor in the collapse.

4. It is believed that inconsistencies and questions remain which cannot be answered by the simple

lateral loads models employed in this preliminary study. Further detailed work should be carried out

involving transient dynamic, inelastic modeling of the Cypress Structure as a whole.

5-20
Figure 5.6.1 Map of northwest Oakland depicting the location of the collapse of 1-880 and bent damage

to 1-980.

5-21
TYPE 1 BENT: 1-880

[Bents 64-69; 82-94; 100-106]

ly I

PILE CAP

PILE FOUNDATION

Section A-A' Section B-B'

#4@ 12" STIRRUPS -*

48"

#4@ 12" - STIRRUPS

722 -# 18

48"

48"

Section C-C

44 -#18

.#4@ 12"

48"

92-96"

5.5"

6.5" T

8" 54"

72" Section D-D'

Typical Box Girder Cell Detail

Figure 5.6.2 Structural configuration of Type 1 bents used on the elevated section of I- 880.

5-22
TYPE 2 BENT: 1-880

[Bent 71 ,72, 75-80 ]

ry-r-,—rp—rp- Cj 1^1 I I 1,1.

I 'l K "1

' ' ' '

PILE CAP

PILE FOUNDATION i.'i ri I.J , ,

Section A-A'

#4 stirrups

@12"

#18

©corners

48"

#4 stirrups

@12"

varies

[42" @bottom ; 54" @top]

Section B-B'

48"

varies

[42" @boltom ; 54" @top]

Section C-C"

[Box Girder Union with Bent Girder]

[prestressing steel not shown]

5.5"

6.5

.8-#4bars

.#4 stirrups

48"

54"

42"

ure 5.6.3 Structural configuration of Type 2 bents used on the elevated section of 1-880.

5-23
TYPE 3 BENT: 1-880

[Bent 70]

r

R/C Girder

Box Girder

I 96"

BA

196"

R/C Girder

> ^ I «

t -J I »

PILE CAP

PILE FOUNDATION

• = Shear Key / Hinge

Vt " l')' "0' V ' V> l "VI

I "J i-si ' 'g II I'si VII) ^ I 1^1 t I I *| rsi < 'g I I I'M

Section A-A'

[typical both sides of bent]

#4 stirrups

@12"

_3

16-#18

Is t t I

48"

varies

[39" @botlom ; 51" @top]

Section B-B'

[Box Girder Union with Bent Girder]

48"

Figure 5.6.4 Structural configuration of Type 3 bents used on the elevated section of 1-880.

5-24
TYPE 4 BENT: 1-880

[BENT 95-98]

# = Shear Key / Hinge

; <. ' .1 I ' 1.1 1^1 1^ I.I 1.1

Section A-A'

[Bents 95, 98]

#4 stirrups

@12" 48"

varies

[36" @bottom ; 48" @top]

Section A-A'

[Bents 96, 97 only]

#4 stirrups

@12"

20.5"

varies

[36" @l>Dltom ; 48" @top]

Figure 5.6.5 Structural configuration of Type 4 bents used on the elevated section of 1-880.

5-25
TYPE 5 BENT: 1-880

[Bent 73, 74 ]

= Shear Key / Hinge

Section A-A'

#4 slirrups @12"

#18 @corners

48"

#4 Stirrups @12"

varies

[48" @bottom ; 60" @lop]

Section B-B'

48"

varies

[48" @bonom ; 60" @top]

Section C-C

[Box Girder Union with Bent Girder]

[includes girder cantilever to east]

Figure 5.6.6 Structural configuration of Type 5 bents used on the elevated section of 1-880.

5-26
16 -#18 Bars

114"

30-#18 bars

#4 ties @1 2" typical

Section A-A' Shear Key / Hinge Detail

4-#10

dowels

1

1

A'

20"

36" 48"

B'

36" 0.5" expansion

joint material

Section B-B' Column Pedestal

4-#4 Stirrups

@ 12"

#4 ties @ 12-inches — 44-#18

60"

Figure 5.6.7 Reinforcement detail at the connection between the upper and lower portal

frames for the Type 1 bents used in Cypress Structure, 1-880.

5-27
Seismic Retrofit for 1-880 : Cypress Structure

Hollow Cell of Box Girder

754" typ.

Southern Box Girder

Northern Box Girder

Seismic Restrainer Cable Assembly

Expansion Joint

Cable

drum

Box Girder

Expansion Joint

V)

]

^ RC bolster

// '7 ^ 0.75-inch cables

mm r/^-i >r in horseshoe configuration

[cable anchors on same side ]

Section A-A'

Box Girder Cell

6" dia. hole

[2 per set ] 9" 9"

View B-B'

Figure 5.6.8 Detail of the Caltrans Phase I Retrofit implemented at the box girder expansion joints

for Cypress Structure, 1-880.

5-28
Figure 5.6.9 Aerial photo of Cypress Structure looking south from the northern

limit of the collapse at 34th Street in Oakland.

5-29
Figure 5.6.10 Aerial photo looking east to the northern limit of the collapsed portion of Cypress

Structure between Bents 110-113.

Figure 5.6.11 Aerial photo looking east to the collapsed portion of Cypress Structure

between Bents 106-108.

5-30
Figure 5.6.12 Aerial photo looking east to the collapsed portion of Cypress Structure

between Bents 104-106.

Figure 5.6.13 Aerial photo looking east to the collapsed portion of Cypress Structure

between Bents 97-104.

5-31
Figure 5.6.14 Aerial photo looking east to the collapsed portion of Cypress Structure

between Bents 94-100.

Figure 5.6.15 Aerial photo looking west to the collapsed portion of Cypress Structure

between Bents 94-100.

5-32
Figure 5.6.16 Aerial photo looking east to the collapsed portion of Cypress Structure

between Bents 86-94.

Figure 5.6.17 Aerial photo looking east to the collapsed portion of Cypress Structure

between Bents 83-90.

5-33
Figure 5.6.18 Aerial photo looking east to the collapsed portion of Cypress Structure

between Bents 78-85.

m

Figure 5.6.19 Aerial photo looking east to the collapsed portion of Cypress Structure

between Bents 71-78.

5-34
Figure 5.6.20 Aerial photo looking east to the collapsed portion of Cypress Structure

between Bents 66-72.

Figure 5.6.21 Aerial photo looking east to the collapsed portion of Cypress Structure

between Bents 62-68.

5-35
Figure 5.6.22 Aerial photo looking east to the southern limit of the collapsed portion of Cypress

Structure between Bents 62-64.

Figure 5.6.23 View of the west side of Cypress Structure looking north between Bents 86-92. These

were all Type 1 bents which demonstrate the uniformity of the collapse mechanism.

5-36
Figure 5.6.24 View of the upper deck of Cypress Structure looking north of Bent 80. This photo

illustrates the negligible lateral displacement of the upper deck relative to the lower deck

during the collapse.

Figure 5.6.25

A view of the west side of Bent

65 of Cypress Structure

illustrating the predominant

failure mode for a Type 1 bent.

The upper frame columns

sheared through the support

pedestal which connected them

to the lower frame. In most

cases the upper columns

remained connected to the upper

deck as shown.

5-37
-.3; m^KS^

Figure 5.6.26

Failure of the west side of Bent 85. In this

particular case a portion of the sheared support

pedestal is still attached to the base of the upper

column. The top of the column failed

explosively and separated from the upper frame

girder.

Figure 5.6.27

Failure of the west side of Bent 90. In this case the supporting pedestal for the upper columns was completely disintegrated; only the interior

face of axial reinforcement that was present in

the pedestal remains imbedded in the lower

column. It can be seen that the failure plane

followed the face of the curved negative moment

steel from the lower frame girder.

5-38
r

Figure 5.6.28

Failure of the west side of Bent 94. The

presence of a central pipe, used to drain the

upper deck, permitted a fracture plane to develop

through the center of the column. The eastern

half of the column disintegrated under

compressive load while the western half spalled

off.

Figure 5.6.29

A view looking north at the west side of Bents

78-80. These were all Type 2 bents which

included prestressing of the upper girders. This

particular mode of failure, in which only the east

columns of the upper frames failed, is noted as a Type A Failure in subsequent discussions.

Type A Failures occurred in Bents 74 through 80

(see figs. 5.6.18 and 5.6.19).

5-39
Figure 5.6.30

A close-up view looking south from Bent 80 at the west side of the upper portion of Cypress

Structure. Note compression spalling along the

interior face of the tops of the columns.

Figure 5.6.31

Due to excessive flexural displacements in the

upper frame girder, most of the post-tensioning

tendons yielded and then fractured. The tendons were unbonded, to permit extension of the top

girder at a later date as part of a planned access ramp. The stored elastic energy drove the

tendons from their ducts, as shown here at Bents

74, 75, and 76 on the east side of Cypress

Structure.

5-40
Figure 5.6.32 There was, apparently, a belief among motorists that it was safer to stop beneath the

bent girders, as shown here, looking south from Bent 80.

Figure 5.6.33

A close-up of the west side of Bent 72 shows

the second, or Type B, variation of failure

observed for Type 2 bents. In this case the

entire double-hinged upper frame column

segment has been ejected from the structure. The 4-#10 bars which passed through the upper

and lower shear keys may be seen projecting

from the base of the upper girder and the top of

the lower column.

5-41
J

Figure 5.6.34 Detail of the base of the two-hinged column that was ejected from the west side of Bent

72. The 4-#10 dowel bars which passed through the center of the shear key have been

pulled free.

Figure 5.6.35 Detail of the bottom of the two-hinged column that was ejected from the west side of Bent

71 showing central pipe used to drain the upper road deck and the 4-#10 dowel bars

used to transfer shear across the hinge.

5-42
\

96

Figure 5.6.36 View looking west towards Bents 96 and 97 which supported the only segment of the

upper deck to remain standing between Bents 62 and 113.

Figure 5.6.37

Detail of the west side of Bent 97 showing

extensive shear cracking at the interior

beam/column joint for the lower frame and

compression spalling at the upper beam/column

joint. Axial reinforcement for the upper frame

columns was continuous with the lower frame

columns on both east and west sides. This, and

the larger cross section, helps to explain why

these Type 4 bents did not collapse.

5-43
Figure 5.6.38 The section of Cypress Structure between Bents 104 and 106 was the only location in

which both upper and lower bent frames failed.

Figure 5.6.39

Close-up of the west side of Bent 105 looking

south, illustrating complete shear failure of the

lower bent beam/column connection.

5-44
Figure 5.6.40

Close-up view of the east side of Bent 108

looking north. This was the last Type 1 bent

prior to the northern limit of the collapse at Bent 113. In addition to the common failure of

the pedestal supporting the upper frame column,

the column of the lower frame sustained a total

shear failure.

Figure 5.6.41 View looking south along the east side of Cypress Structure from Bent 115. The northern

limit of the collapse extended to within six feet of Bent 113.

5-45
Mode 1: 2.5 Hz Mode 2: 6.4 Hz

Mode 5: 23.4 Hz

Figure 5.6.42 Vibrational mode shapes for Type 1 bents used in the Cypress Structure, 1-880.

5-46
Mode 1: 2.0 Hz Mode 2: 6.3 Hz

Mode 5: 23.3 Hz

Figure 5.6.43 Vibrational mode shapes for Type 2 bents used in the Cypress Structure, 1-880.

5-47
TYPE 1 BENT: 1-880

Pents 64-69; 82-94; 100-106]

' ' '/• V

Stress Trajectory

Detail

Figure 5.6.44(a) Type 1 bent.

2F = 633 kips

F = 316 kips

Type 1 Bent

21 kips/foot W I I I II II

Figure 5.6.44(b) Displaced shape of Type 1 bent under applied loads showing zones of principal tensile

stress sufficient to induce cracking.

5-48

I
c

o

Jid CD

M—

C3L,

onO.

O

vo

u

a

d

s

CO

c

« . ^ o

\ \ I /

7

CO

-•—

CO

. CD

CO "o

CO 0

LU Q.

O

0

T3

D)

CD

E

CO

CD

o

/

+

X

X

X

/

/

/

I /

^ 1 )

X

X

X

X

X

X

X / /

X X X

X X X X

X X X X A- A- -1- X X X X \ \

X X X X \ \ X

X X X \ X \

X X X XX

c

<D E

o o

-J o

c o

.2 «

- ^

<U to

,0 W)

O -a 5 ^

O TO

I— li — T3

TO O

c

'-3

to

9

5-49
TYPE 2 BENT: 1-880

[Bent 71,72, 75-80]

T-i—p-^—m—-nl'""! I .1 1 J I"" I

I. I 1-^, I A I -t

Figure 5.6.45(a) Type 2 bent.

stress Trajectory

Detail

2F = 633 kips

F = 316 kips

Type 2 Bent

21 kips/foot

21 kips/foot

\ \ \ \ WW

Shaded areas indicate zones where stress exceeds tensile strength of concrete

Figure 5.6.45(b) Displaced shape of Type 2 bent under applied loads showing zones of principal tensile

stress sufficient to induce cracking.

5-50
co

.ti CO

ft

X

X

y

X

X

X

X

X

X

X

X

X

X

X

X

X

X

X

X

X

AX

X

X

X

X

X

X

X\

\

X

X

\

\

\ > V

X

X

X

\\ ^ / / / cE

_2oo0o

s XX

"A

s

(/2 a, os

C/5C

(S

o

c

o

.2 ^ BO 7^ o «

on ^ m eg

"c '-a

Cu o- 3

GO

5-51
TYPE 5 BENT: 1-880

[Bent 73, 74 ]

Stress Trajectory

Detail

—r-i—n:T j"r; i'i''r'i "

i" i "-i" '^ 'SI K I ' I '

Figure 5.6.46(a) Type 5 bent.

Type 5 Bent

21 kips/foot U W i I I I I ;

2F = 633 kips

21 kips/foot

\ \ \ \ \ \ \ \

F = 316 kips

Shaded areas indicate zones where stress exceeds tensile strength of concrete

Figure 5.6.46(b) Displaced shape of Type 5 bent under applied loads showing zones of principal tensile

stress sufficient to induce cracking.

5-52
T3

X Jr -f

^ X X

AX

X

X

>< \

X \ \ k

^ ^ ^ J \ \ ^

5-53
TYPE 1 BENT: 1-880

(Bents 64-69; 82-94; 100-1061

First Cycle West

Initiates cracking of pedestal

First Cycle East

Flexural cracking at interior of joint

[2) First Cycle West

flexural cracking

First Cycle East

initial cracking of pedestal

East

1 1 II Ivl I'l f

,

A: Crack Initiation Under Lateral Load

si

1

Tops of many columns

exhibit explosive compression

failures in some cases column segment separates and falls

Base of column sfiears free often taking with it a portion

of column pedestal

'i;i\Y"i" i "r4i"i'j

'

I J I J i^i rill KIM I I I 1 i-i

B: Typical Failure Mode for Type 1 Bents

Figure 5.6.47 Top: Anticipated crack initiation sequence for Type 1 bents under lateral loading.

Bottom: Observed failure mode for Type 1 bents at Cypress Structure.

5-54
TYPE 2 BENT: 1-880

[Bent 71,72, 75-80]

Rrst CydG East Rexural aacking

at interior comer

of joint

Rrst Cyde West

Rexural aacking

at interior corner

of joint plus reverse shear aacks in joint

First Cyde East

1 J

Initial sheaf aacking at joint plus flexural cracking of girder

EAST

Crack initiation Under Lateral Load

Failure Mode Type B

Figure 5.6.48 Top: Anticipated crack initiation sequence for Type 2 bents under lateral loading.

Middle: Observed failure mode "A" for Type 2 bents at Cypress Structure.

Bottom: Observed failure mode "B" for Type 2 bents at Cypress Structure.

5-55
TYPE 5 BENT: 1-880

[Bent 73, 74 ]

First Cycle East

Flexural and shear

cracks initiate in joint

First Cycle West

Flexure cracking

initiates in negative

moment region of cantilever

A: Crack Initiation Under Lateral Load

1^^

; i "i.' Vi" i!'i "

( "J

' >

»'

Due to lack of joint confinement and insufficient penetration of girder

negative moment steel into the joint, column segment

shears free following formation of flexural cracks

B: Observed Collapse Mode for Bent 74

Figure 5.6.49 Top: Anticipated crack initiation sequence for Type 5 bents under lateral loading.

Bottom: Observed failure mode for Type 5 bents at Cypress Structure.

5-56
1-880: 3-SPAN UNIT

Figure 5.6.50 Vertical vibration mode shape for Cypress Structure. Period = 0.16 seconds (6.2 Hz).

Boundary conditions at the ends of the spans were chosen such that the effective model

included nine spans of the structure; only the central three are shown here for clarity with

the termination points for the box girders occurring at the expansion joints.

5-57
5.7 Bay Bridge

The San Francisco-Oakland Bay Bridge is a part of Interstate 80 and is the largest engineered structure

within the zone of influence of the earthquake. It is a key lifeline element which significantly affects traffic

flow in the Bay area. The bridge spans 8.3 miles (13.3 km) and consists of a series of interconnected

structural systems. The two longest spans each measure 2,310 feet (704 m) and are the central sections

of the twin suspension bridges which comprise the west bay section from the San Francisco wharf to Yerba

Buena Island on the east (see figs. 5.6.1 and 5.6.2). From the east end of the suspension spans the fivelane double-deck highway leads through the south end of Yerba Buena Island via the world's largest

single bore tunnel and empties into the east bay segment which is comprised of a series 22 truss spans

of various configurations which lead 1.75 miles (2.8 km) to Pier E23 in the Oakland mud flats, west of

the toll house and administration building (see figs. 5.6.2 and 5.6.3.). The truss segments which support

the roadway generally consist of upper and lower 54 foot (16 m) wide concrete decks supported by

longitudinal girders resting on transverse plate girders. The transverse girders connect the deck systems to

the two parallel trusses as shown in figure 5.6.6.

There was no apparent damage as a result of the earthquake to the west bay suspension bridges. First

reports of the earthquake via the media, however, showed the dramatic collapse of the 50 foot (15 m)

link spans at pier E-9 (see fig. 5.6.4), approximately halfway between Yerba Buena Island and the Oakland

landing. Inspection revealed that both the upper and lower decks had simultaneously been pulled off

their bearing seats as the truss east of pier E-9 moved, by Caltrans measurements, as much as 10 inches

(250 mm) east during the first excursion of motion before returning 5 inches (130 mm) to the west, leaving

a residual 5 inch (130 mm) eastward displacement. The western girder seats for the upper and lower

bridge decks (see fig. 5.6.6) were simply supported configurations in which the bottom flange of the deck

girders sat on a riveted angle bracket while two vertical angle brackets constrained movement of the

girder web. The bearing seat was only 6 inches (150 mm) wide, so that when truss E9-E10 moved

eastward both top and bottom decks slipped off their respective angle brackets and pivoted downward,

landing as shown in figures 5.6.4 and 5.6.5. This mechanism is evident when viewing figures 5.6.7 and 5.6.8

which show closeup photos of the western bearing seats for the lower deck girders and the eastern hinge

seat for the upper deck girders, respectively. There is practically no damage visible in figure 5.6.7 to the

bearing seats, indicating that the girder flanges had cleared the edge quickly and fell away without any

further impacting before the lower deck girders landed on a transformer station mounted on the utility

deck on the north side of the bridge. The eastern deck girder supports for the link spans, which were in

contrast bolted to the girder, were severely distorted (fig. 5.6.8) and in some cases torn loose. It has been

suggested that were it not for the presence of the transformer station both spans would likely have torn

5-58
free from the girder restraining bolts and fallen into pier E-9, causing far greater damage*. During the

fall the upper deck cut through several power and communication lines and additionally severed a water

main on the south side, but otherwise the damage was confined to the decks themselves.

A few comments are in order to understand the seismic displacements observed in the east bay trusses.

Although there are 22 total spans involved, 16 of these are grouped into four structural units with regard

to longitudinal (generally temperature induced) displacements. Figure 5.6.3 indicates that the first of these

units consists of the 2420 foot (738 m) cantilevered truss section connecting pier El, the final pier on

Yerba Buena Island, to pier E-4. The E-1 detail is a fixed connection; the pier E-4 connection allows for

expansion. Caltrans has reported indications (paint scrapings) which show that pier E-4 experienced as much as 12 inches (305 mm) of overall movement without damage to the unit. The expansion joint was

designed to accommodate 24 inches (610 mm) of displacement at this particular location. Unit 2 consists

of the five 509 foot (155 m) truss spans connecting pier E-4 to pier E-9. No damage was reported to

this unit.

Pier E-9 is a four-column diagonally braced tower which was designed as a laterally stiff mid-point support

for the east bay spans. Its two western column tops form the seats for the truss shoes of the 509 foot

span from pier E-8 to pier E-9 and are bolted to these shoes. Likewise, the eastern column tops were

connected to the western truss shoes of span E-9 to E-10. These shoes were connected to pier E-9 by

means of 20 1 inch (25 mm) bolts each. During the earthquake Unit 3, which also includes span E-10

to E-11, moved eastward as previously described before returning some 5 inches (130 mm). All 40 bolts

in the E-9 connecting shoes appeared to be failed in both tension and shear. This is explained by the

fact that the shoe detail consists of a 12 inch (305 mm) pin connection between the truss and the shoe,

with the centerline of the pin lying at a level 27 inches (690 mm) above the base of the shoe. Lateral

displacements of Unit 3 to the east would therefore, in addition to applying substantial shearing forces to

the 40 bolts, would also apply a significant overturning moment that would largely be resisted by the

western-most bolts in tension. Fortunately the shoe seat was sufficiently broad that the shoe remained

atop the seat. Caltrans reported that the northeast shoe had a permanent displacement of 5.5 inches (140

mm) east and 5/8 inch (16 mm) north following the quake; the southeast shoe moved 5 inches (130 mm)

east and 1/2 inch (13 mm) north. These were subsequently forced back into place with a jacking load of

400 tons (3560 kN). It was reported that this load decreased as fiiction at the truss shoes was overcome

^Governor's Board of Inquiry Meeting, State of California: "The 1989 Loma Pricta Earthquake,"

December 13-14, 1989, Hyatt Regency International Hotel, Oakland, CA

5-59
and that there was some elastic rebound from pier E-10^. Within 4 weeks of the quake, Caltrans had

replaced the deck girders for the pier E-9 link spans, placed precast concrete panels between them, cast

the composite deck, and reopened the bridge for traffic. The 40 1 inch (25 mm) bolts connecting the

E-9/E-10 truss shoes to the eastern column tops of pier E-9 were replaced with high strength bolts.

Whether this is to the advantage of the bridge during future earthquakes is subject to review. Opinions

have been expressed by some^ that the shearing of these bolts served as a fuse, isolating pier E-9 from

otherwise large lateral loads which could have led to disastrous column buckling.

Subsequent detailed studies on a pier by pier basis carried out by Caltrans revealed no damage to the six

292 foot (89 m) trusses and their supporting piers which comprise Unit 4. However, all of the simply

supported 292 foot (89 m) trusses east of pier E-17 had all 12 1 inch (25 mm) diameter retainer bolts

broken at each shoe location. All north and south movements were restricted by keeper plates placed next

to the bearing shoes and attached to the piers. The bolts which failed on these truss shoes also appeared

to have been due to combined tension and shear. This shear failure permitted the various trusses between

pier E-17 to E-23 to move somewhat independently. The expansion side of each truss was clamped by

earthquake restrainers installed in 1975 as part of the Phase I retrofit program. The restrainers all

sustained some form of damage but kept the expansion bearing in place. The final resting place of the

north and south shoes are, according to Caltrans testimony to the Governor's Board of Inquiry, as follows:

Pier E-17: no apparent movement

Pier E-18: north shoe 3/4" (19 mm)west; south shoe 1/8" (3 mm) west

Pier E-19: north shoe 1" (25 mm)west; south shoe 1.5" (38 mm) west

Pier E-20: north shoe 2" (51 mm)west; south shoe 2" (51 mm) west

Pier E-21: north shoe 2.25" (57 mm)west; south shoe 2 5/8" (67 mm) west

Pier E-22: north shoe 5" (127 mm)west; south shoe 2.5 "(64 mm) west.

On the basis of these displacements, as well as paint scrapings on the truss supports, it was concluded that

piers E-18 through E-22 moved as much as 10 inches (250 mm) in the east-west direction before coming

to rest.

A final failure occurred in the span just east of the massive concrete structure at pier E-23. The rivets

connecting the floor beam to the concrete pier were sheared off, permitting span E-23/E-24 to move

^Governor's Board of Inquiry Meeting, State of California: "The 1989 Loma Prieta Earthquake,"

December 13-14, 1989, Hyatt Regency International Hotel, Oakland, CA.

^Earthquake Engineering Research Institute, EERI 1989, EERI-NRC-NCEER Congressional Briefing,

"Loma Prieta (Northern California) Earthquake," Senate Russell Office Building, Room 253, November

30, 1989.

5-60
eastward. The floor beam on the east side of pier E-23 moved, according to Caltrans measurements, 2.5

inches (640 mm) in both the north-south and east-west directions, with a residual displacement of 2 inches

(51 mm) east and 1.5 inches (38 mm) north of its original position. The floor beam pulled the attached

stringers to within one-quarter inch (6 mm) of the edge of the beam seats. The webs had already moved

outside the guide angle (see Detail A, fig. 5.6.6). Had the displacement of the floor beam been a quarter

inch further this span, which was similar to the link span at pier E-9, would also have collapsed. The

construction of extended beam seats for this span contributed to the delay in the subsequent re-opening

of the Bay Bridge. It is noted that between pier E-9 and pier E-23 there was a residual opening

displacement following the quake.

Summary

1. Damage to the San Francisco-Oakland Bay Bridge was concentrated within three general areas.

These included the collapse of the upper and lower link spans at pier E-9 and the shearing of the

truss shoe retainer bolts for span E-9/E-10; the general shearing of truss shoe retainer bolts for spans

between piers E-17 and E-23; and the near loss of span E-23/E-24.

2. In general, the entire structure performed remarkably well and what damage there was proved to be

minor. It may be useful, however, to conduct a detailed investigation of the subsoil conditions,

particularly beneath the east bay spans, characterize their dynamic properties, and then use this as a

set of boundary conditions for a full three-dimensional transient dynamic analysis of the complete

section from Yerba Buena Island to pier E-23 to arrive at anticipated maximum displacement demand

at the truss shoe connections. These data would then permit better design of the shoe size, as well

as the required strength of the restraining bolts.

5-61
Figure 5.7.1 View looking east from San Francisco to Yerba Buena Island showing the two main

suspension spans for the Bay Bridge. No damage was detected in this section.

5-62
.2?

5-63
YB1 YB2 YB3 YB4 E1

5Z

Pin

-512'-

E3 E4

UNIT1

Fixed

I UNIT 2

50' 290' 292' 292' 291' 291' 291' 291' 292^

Fixed \ei Rxed

E9 E10 E11 E12 E13 E14 E15 E16 E17

\ UNITS UNIT 4

E17 El 8 E19 E20 E21 E22 E23 'E24

Figure 5.7.3 Detail of the east bay truss spans. Unit 3 was determined to have moved eastward as much as 10 inches, causing the collapse of the two deck segments at Pier E-9. Additional

damage, generally involving the failure of retainer bolts, occurred between Piers E-17 and

E-24.

5-64
Figure 5.7.4 Aerial photo looking north at Pier E-9 on October 19, 1989 showing the collapsed upper

and lower 50 foot deck spans.

Figure 5.7.5 Schematic detail of Pier E-9 looking north. Evidence indicates that the 290 foot truss segment on the east side of the pier moved eastward, simultaneously unseating both Upper

and lower deck segments.

5-65
San Francisco - Oakland Bay Bridge

[ Typical Deck Cross Section at Pier E-9 ]

1 6' typ.

South Truss

58'

Upper Road Deck

Detail A

[ Deck Girder Seat ]

North Truss

Lower Road Deck

mTT

South Truss Shoe North Truss Shoe

38'

_ J.

Main Transverse Girder

Upper West

Transverse Girder

Upper Deck Girder

Detail A : Deck Girder Seat

6-inch Seat for Deck Girder

Figure 5.7.6 Cross-sectional detail of the Bay Bridge at Pier E-9, showing upper and lower roadway

transverse support girders and longitudinal deck girder seats. The connection of the

longitudinal girders to the transverse girders consisted of a simple bearing seat on the

west side; The eastern beam seat was bolted.

5-66
Figure 5.7.7 Close-up of the western girder seats for the lower road deck at Pier E-9. Lack of

significant damage to these seats, and to those supporting the upper road deck, indicates

that both upper and lower segments moved eastward by an amount sufficient to permit

both to slide off the western seats and drop simultaneously.

Figure 5.7.8

Close-up of the eastern seat for the upper deck

girder showing distortion of bolted beam seat.

5-67
Figure 5.7.9 Close-up of the northeast shoe for the 290 foot truss between Pier E-9 and E-10. During

the earthquake this truss moved eastward as much as 10 inches before returning to leave a 5.5-inch residual displacement to the east and 5/8-inch to the north.

5-68
5.8 Struve Slough Bridges

The team did not visit the Struve Slough bridges, located on state Highway 1 south of Watsonville, but

this particular collapse has been described in detail elsewhere (Civil Engineering, 1989). At the California

Governor's Board of Inquiry meeting of December 14, 1989, Caltrans officials reported the following:

"The Struve Slough Bridges were a pair of reinforced concrete T-beam bridges built in 1964. The

bridges are not unusual and were considered typical of bridges built at that time. Approximately 800

feet (244 m) long and 34 feet (10 m) wide, the structures were founded on pile extensions at each

bent and on monolithic diaphragm abutments. The spans were 37 feet (11m) long and the pile bents

consisted of 4 driven piles each. The piles were about 80 feet (24 m) long and were driven full length

and extended to the superstructure. Three expansion hinges were located in each bridge in spans 6,

11, and 16 on the west bridge and spans 6, 11, and 17 on the east bridge."

"Seismic retrofit consisting of the addition of cable restrainer units to each expansion hinge was added

to the bridges in 1984. About 7 spans of the eastern structure and 10 spans of the western structure

collapsed. Collapse was generally downward, however the western bridge moved transversely about

2 feet (610 mm). The structure shows little evidence of significant forces [being transmitted] to the

abutments or superstructure. The hinges showed no evidence of banging and the abutments did not

indicate that any movement took place there. The approach fills settled about 3 inches (75 mm)."

"The hinge restrainers performed well and held the structure together in spite of the large downward

movement. Some evidence existed at the site that auto traffic had crossed the bridge during or

immediately after the event. The pile extensions under the structure were lightly confined with #3 wire

at about 12 inch (305 mm) pitch and failed primarily at the top. Some piles apparently also failed

some distance below the ground surface. The soil displacement at the base of several of the piles was

in several directions up to a maximum of 18 inches (460 mm). The eastern structure dropped about

5 feet (1.52 m) onto damaged pile extensions. The western structure dropped to the ground surface

about 8 to 10 feet below (2.44 to 3.05 m). A few of the piles on the western bridge separated from

the bent, displaced, and punched through the deck slab as the superstructure fell to the ground (see

fig. 5.8.1)."

5-69
View looking north of State Highway 1 crossing Struve Slough, approximately 1 mile south

of Watsonville. Piles on the left bridge separated from the bent, displaced, and punched

through the deck slab as the superstructure fell to the ground (photo courtesy EQE

Engineering).

5-70
5.9 Interstate 280

Aside from the damage to 1-880 and to the San Francisco-Oakland Bay Bridge, structural damage at four

other locations was investigated. These locations are shown in figure 5.9.1, and the nature of the damage

are described in this section and the sections to follow.

Interstate 280 (1-280) traverses the southern part of San Francisco from west to east and proceeds in a

northern direction along the eastern shoreline. The highway was named the Southern Freeway when it was designed. The portion south of Army Street was designed according to the 1961 AASHO

Specifications (AASHO, 1961), and the portion near Sixth Street was designed according to the 1965

AASHO Specifications (AASHO, 1965). Damage was observed at two locations: several blocks south of

Army Street and at the Sixth Street ramp (see fig. 5.9.1).

Figure 5.9.2 is an aerial view, looking south, of 1-280 south of Army Street. At this location a twolevel elevated highway (to the south) transforms into to a single level divided highway (to the north). In

addition, there is an exit ramp for the lower level northbound section. Damage was observed at Bent

numbers 48, 51, and 52.

Figure 5.9.3 is a schematic elevation view (looking north) of Bent 48. The top girder is post-tensioned

and supports at the columns are designed as pinned connections, with details similar to those used on I880. The column reinforcement includes #18 longitudinal bars and #4 ties at 12 in. (0.30 m) spacing.

The damage occurred at the top of the west-side column. Figure 5.9.4 shows the nature of the damage,

which appears to be a combination of diagonal tension failure and splitting parallel to the longitudinal

column steel. Note that the girder is permanently displaced toward the east.

Figure 5.9.5 is a schematic elevation view of Bent 52. The configuration of Bent 51 is similar. The

upper, post-tensioned girder is monolithic with the column on the west side and is supported as a pinned

connection on the east side. Figure 5.9.6 is view (looking south) as seen from the lower northbound lanes;

the upper girder of Bent 52 is in clear view. Damage occurred to the east-side columns of Bents 51 and

52 at approximately the elevation of the roadway. Figure 5.9.7 shows the damage in Bent 52 as it

appeared on the northern column face. The damage in Bent 51 was similar as can be seen in figure

5.9.8, which shows that southern column face. Figure 5.9.9 is a close-up view of the damage zone in Bent

51. It is seen that the concrete is extensively fractured and the #4 ties lost their effectiveness because

of the inadequate anchorage provided by the 90-degree hooks. The corner #18 bar also appears to have

buckled.

5-71
Some minor damage occurred due to pounding. It is seen in figure 5.9.2 that where the roadways become

side-by-side, the columns supporting the upper roadway are close to the lower roadway. One of these

columns which support the upper roadway was damaged due to pounding with the lower roadway, as can

be seen in figure 5.9.10.

The complex geometry of this portion of 1-280 is likely to be a factor in explaining the causes of the

damage. Figure 5.9.11 is a schematic to illustrate the changing structural configuration as the highway

changes from a two-level elevated structure to a single-level divided structure. In part (c) of the figure,

a tall structure is connected to a short structure. The natural period of vibration of the short structure

is shorter than the natural period of the tall structure. Hence it is likely that at some time during the

shaking, the roadways were moving in opposite directions. This would lead to high lateral forces at the

junction of the two structures, which is consistent with the location of the observed damage. Where the

two roadways are completely separated, as in figure 5.9.11(d), motion in opposite directions would lead

to pounding as was observed.

The other damage to 1-280 was observed at the Sixth Street ramp, where the highway crosses China Basin.

Figure 5.9.12 is an aerial view of the site showing the elevated exit ramp crossing over the main highway.

Figure 5.9.13 is a ground-level view (looking west) showing the exit ramp above the northbound lane of

1-280 exit. The photograph was taken several weeks after the earthquake, and wooden cribbing was being

positioned to provide temporary support. The exit ramp is typically supported by a single pier except at

this location, where it is supported by a bent (number N-35). The supporting structure for the highlevel ramp is attached to the lower highway in a manner similar to that shown in figure 5.9.11(c), i.e., one

column is part of the bent for the main roadway, and the other column is free-standing. Damage occurred

in the girder supporting the ramp. Figures 5.9.14(a) and 5.9.14(b) show the nature of the damage on the

east side and west side of the bent, respectively. On the east-side, diagonal cracks developed in the girder

with a major crack running diagonally across the corner. On the west side, many cracks developed toward

the corner.

The other place of observed damage at the Sixth Street exit was at Bent 32 of the main highway. Figure

5.9.15(a) shows the appearance of this portion of the structure. The roadway is placed unsymmetrically

with respect to the bent. On the east side, a portion of the bent extends beyond the boundary of the

roadway. Figure 5.9.15(b) is a close-up view of the girder showing the series of diagonal cracks that

developed.

5-72
r

The damage to these portions of 1-280 is likely the result of several factors: the structures were designed

before modern seismic provisions had been adopted; they were located in areas of deep soil deposits which

probably experienced large ground displacements; and the structures had complex geometries in which

components with different natural frequencies were joined. To arrive at adequate methods for repair and

strengthening, it will be necessary to perform dynamic analyses, using realistic models, of these structures

to arrive at realistic estimates of the forces and displacements that might occur in future earthquakes.

Figure 5.9.1 Location of damage to elevated highway structures in San Francisco.

5-73
Figure 5.9.2 Aerial view of 1-280 south of Army St., note the transition from a two-level elevated

structure to a single-level, divided structure (view is looking south).

5-74
BENT 48

Roadway

10' X

5'

6'x5' P/T Girder

4' X 5' —

9' x5' I II II ^1—If

R/C Girder

6'x5'

\ \ \ \ \ "s X \ \ N *^ \

S S S \ \ \

East

/ f / / f

k \ \ X X X \ \ \ \ s f ^ f f ^ f

w \ *% \

f f f . \ \ s

y / / / / / y / / ^ s s X X

r( \ s s \ X

f f f f y y / X X

/ y f

X X X X

/ / f f f / f

X X

X X

y y f f = Shear Key / Hinge

Figure 5.9.3 Schematic elevation view of Bent 48 of 1-280 (looking north).

Figure 5.9.4 Damage to west-side column of Bent 48 (looking toward south).

5-75
BENT 52

Roadway

6' X 4.5'

^4.

I II 11 lIZZlC—

1

10' X 4.5'

P/T Girder

7^

4'x4.5'

5* X 4.5'

6' X 4.5'

^ ^ ^ ^ ^

\ \ \ \ \ y y y y y y J

\ N \ X N

y y y y y y y X N *\ \ '\ \ '\ y y y y y y y

. N N \ \ N N \

y y y y y y y

, \ \ x \ \ \ \

y y X

y . s y

Shear Key / Hinge

Figure 5.9.5 Schematic elevation view of Bent 52 of 1-280 (looking north).

Figure 5.9.6 View looking south on 1-280 showing the upper girder of Bent 52.

5-76
Figure 5.9.8 Diagonal cracking of the east-side column of Bent 51 (south face).

5-77
Figure 5.9.10 Pounding damage at 1-280 south of Army Street.

5-78
(a) (b)

/ X X ^ ^ /' / / / /

(c) (d)

^^^1 / / 1' / / / /

\ \ s \ \ \ \ ^^^7 / -f ^ f f -f f

\ \ \ \ \ \ x \ \ \

Figure 5.9.11 Schematic illustration of the various structural configurations as 1-280 undergoes transition

from a two-level to a single-level elevated highway.

5-79
Figure 5.9.12 Aerial view of the Sixth St. exit of 1-280.

Figure 5.9.13 The Sixth St. exit ramp where it passes over the northbound lanes of 1-280. (View is toward west; cribbing was added to provide temporary support.)

5-80

Figure 5.9.15 (a) Bent 32 supporting 1-280 at Sixth St. exit ramp and (b) close-up of girder showing

diagonal cracking.

5-82
5.10 Embarcadero Freeway (1-480)

The Embarcadero Freeway (1-480) is a two-level elevated highway which provides the Financial District

with access to the San Francisco-Oakland Bay Bridge. According to the as-built drawings supplied by

Caltrans, the freeway was designed according to the 1953 AASHO Specifications (AASHO, 1953). Figure

5.10.1 is an aerial view of the portion of 1-480 north of Mission Street. The historic Ferry Building is east

of the freeway.

Figure 5.10.2 is a ground-level view looking north along the west side of the freeway; Bent 76 is in the

foreground. The vertical metal covers which can be seen at the top girder-column joints protect restraining

tendons which were added during the earthquake upgrading in 1972. Figure 5.10.3 is a ground-level view

from the east side looking north; Bent 78 is in the foreground. Mission Street is located between Bents

78 and 79 (numbers increase toward the north). The observed damage occurred at the lower girdercolumn joints in bents located north and south of Mission Street.

Figure 5.10.4 shows the configurations of the bents to the north and south of Mission Street. As can be

seen at the bottom of the aerial photograph in figure 5.10.1, the Embarcadero Freeway undergoes a

transition in width south of Mission Street, and this is the reason for the different bent configurations.

For the wide portion of the freeway, the upper girder is post-tensioned and supported on pinned joints.

A detail similar to that used in the design of 1-880 was used for the pinned joints. Likewise, for bents

with post-tensioned girders, column segments were pinned at both ends on one of side of the bent. The

predominant damage was in the form diagonal cracking within the lower girder-column joint, as indicated

by the heavy lines in figure 5.10.4. This indicates that the weakness of the Embarcadero Freeway is similar

to that of the collapsed portion of 1-880. In no cases were the cracks as severe as those observed in the

columns of 1-280.

Figure 5.10.5(a) is a close-up view of the damaged joint on the west side of Bent 78. Cracking was

observed in the joints at both sides of the bent. Diagonal cracks in the joint were in the opposite

direction to those at other joints on the west side of the bents. On the south face of the column in figure

5.10.5(a), the cracking was more severe and some of the concrete cover had spalled. Figure 5.10.5(b) is

a close-up view of the joint of Bent 79 on the east side. This was the most severe damage observed on

the east side of the freeway.

Cracks were observed in the asphalt pavements and concrete sidewalks beneath the freeway, indicaiing that

there were large ground movements during the earthquake. For example, figure 5.10.6 shows a cracked

5-83
concrete slab at the base of Bent 77 on the east side. The scrape marks on the column are a permanent

record of the amount of relative movement that occurred during the earthquake. It also appears that the

ground has settled about 1 inch (25 mm).

The reasons for the damage to the Embarcadero Freeway are likely the same as the reasons for the

damage to 1-280. The structure was built prior to the adoption of modern seismic design criteria, and the

structure is located in an area which experienced large ground displacements. While not as obvious as

in 1-280, the damage occurred in that portion of the Embarcadero Freeway where a transition in the bent

configurations occurred. Thus dynamic analyses, using accurate structural models, are needed to determine

the response of the structure for earthquake loading.

Figure 5.10.1 Aerial view of the Embarcadero Freeway north of Mission St.

5-84
Figure 5.10.2 West side of Embarcadero Freeway south of Mission St., Bent 76 is in foreground.

Figure 5.10.3 East side of Embarcadero Freeway at Mission St., Bent 78 is in foreground.

5-85
BENT 76 BENT 77

9 ,5'x 4'

P/T Glrdoi

4.5' X 4'

jLJLjgc^ai:

P/T Girder

4.5' X 4' 'CZll ^l_JI ^1 II ^1 Shear Key / Hinge / 9'x 4'

BENT 78 BENT 79

5'x 6'

P/T Girder

9 5'x 5'

R/C Girde

CZ] CZl [=CZl dD

:9.5'x4.5':

[J I—^1

\nz\\ II iLUi ^1 I

9.5' X 4.5'

BENT 80

R/C Girder

4' X 4.5'

9' X 4 5' \

Figure 5.10.4 Schematic of bent configurations and damage at lower girder-column joints.

5-86
I

J

Figure 5.10.5 (a) Close-up view of west side of Bent 78 showing diagonal cracking in joint and vertical

cracking at girder-column boundary; (b) close-up view of east side of Bent 79 showing

diagonal cracking in joint.

5-87
Figure 5.10.6 Evidence of large ground movement at east side base of Bent 77.

5-88
5.11 U.S. Highway 101

The northernmost portion of U.S. 101 (originally named as the Central Viaduct) was designed according

to the 1953 AASHO Specifications (AASHO, 1953). This was the last section of the highway to be built

and extends from South Van Ness Avenue to Turk Street. At the southern end, U.S 101 is a divided,

elevated highway. It undergoes a transition to a two-level elevated highway as its direction changes from

east-west to north-south. At the southern end near Van Ness Avenue, the concrete roadway is supported

by a steel frame. From north of Mission street, the roadway is supported by concrete structures.

One block south of Hayes Street, the highway becomes two lanes in each direction. At Bent 40, the

roadway widens to accommodate additional lanes in the future. Damage was observed in the widened,

elevated section at Bents 42 and 43, which are located north of Hayes Street. These are shown in figure

5.11.1, a view of the east side of the highway. Figure 5.11.2 is a schematic of the configuration of Bent

43 where serious damage was observed. The configuration of Bent 42 is similar to Bent 43 except that

the bent is skewed to the direction of the roadway. The design was similar to the other elevated highways

built in the mid 50's. The post-tensioned girders are supported on columns with pinned connections.

Figure 5.11.3(a) is close-up view of the east-side column of Bent 42. A series of diagonal cracks occurred

at the middle of the column, and finer cracks developed at the top of the column inclined in the opposite

direction. The more serious damage was observed at Bent 43. As shown in figure 5.11.3(b), extensive

diagonal cracking developed just above the roadway.

The damaged portion of U.S. 101 is similar in design to the other damaged highway structures that have

been discussed. The minimal transverse column reinforcement provided little shear strength for the

columns. Further analysis is needed to explain why failure was concentrated within the columns rather

than the girder-column joint, as was observed in other damaged elevated highways of similar design.

5-89
Figure 5.11.1 View of east side U.S. 101 north of Hayes St. showing damaged Bents 42 and 43.

BENT 43

4' X 4'

5'X4'

P/T Girder

][![

7,5' X 4'

R/C Girder

7"

II II II

9'X4'

ift

\ •% N \ \ \ \ \ . \ s \ s \ \ X

X X /i::::r j I /I f > / f / / / / / / / I | ,1 * [:i x / // ff/f / V -fxl' I::! .•(:::}' V X

s \ :: s 's 's X N X s \ \ \ W ^ XXXXXXNXXX F::; i:;4 X X X

/ X A-A ts: X v xxyxxx/X/xl / {••:] f.f/f^ffi'.f / X:-: / .(::}' X X x(::> XXXXXXXXXX'-I^ vkxxxxxxxxxx t:;: \A XXX

• = Shear Key / Hinge

Figure 5.11.2 Schematic of Bent 43, which had the more severe damage.

5-90
Figure 5.11.3 Close-up views of east side columns of (a) Bent 42 and (b) Bent 43

5-91
5.12 References

AASHO, 1953, "Bridge Design Specifications Manual," Standard Specifications for Highway Bridges,

American Association of State Highway Officials, Washington, D.C.

AASHO, 1958-59, "Interim Specifications, Bridges 1950-1959," Subcommittee on Bridges and Structures,

American Association of State Highway Officials, Washington, D.C.

AASHO, 1961, "Bridge Design Specifications Manual," Standard Specifications for Highway Bridges,

American Association of State Highway Officials, Washington, D.C.

AASHO, 1965, "Bridge Design Specifications Manual," Standard Specifications for Highway Bridges,

American Association of State Highway Officials, Washington, D.C.

AASHTO, 1977, "Bridge Design Specification Manual," Standard Specifications for Highway Bridges,

American Association of State Highway and Transportation Officials, 12th Edition.

ACI, 1989, "Building Code Requirements for Reinforced Concrete (ACI 318-83)," American Concrete

Institute, Detroit, MI.

ATC 1978, "Tentative Provisions for the Development of Seismic Regulations for Buildings," Applied

Technology Council, Palo Alto, CA.

Bertero, V., et al., 1989, "The Collapse of 1-880 as a Result of the October 17, 1989 Loma Prieta

Earthquake," University of California at Berkeley Research Report #89-14.

Caltrans, 1949, "Caltrans Bridge Department Design Supplement for Earthquake Loads, 1949 Edition."

California Department of Transportation, Sacramento, CA.

CDMG, 1989a, "A Quick Report on CSMIP Strong-Motion Records from the October 17, 1989 Earthquake

in the Santa Cruz Mountains," State of California, Department of Conservation, Division of Mines and

Geology, Office of Strong Motion Studies.

CDMG, 1989b, "Strong Ground Shaking from the Loma Prieta Earthquake of October 17, 1989 and its Relation to Near-Surface Geology in the Oakland Area," A Preliminary Report to the Governor's Board

of Inquiry on the 1989 Loma Prieta Earthquake. California Department of Conservation, Division of Mines

and Geology.

Civil Engineering, 1989, "The Great Quake: On-Site Reports," Civil Engineering, December, 1989, American

Society of Civil Engineers (ASCE), New York, NY, pp 44-47.

esse, 1988, "California at Risk: Reducing Earthquake Hazards 1987-1992," A Report of the California

Seismic Safety Commission, Report No. SSC 88-02.

EERC, 1989, "Preliminary Report on the Seismological and Engineering Aspects of the October 17, 1989

Santa Cruz (Loma Prieta) Earthquake," EERC, College of Engineering, University of California at Berkeley, Report No. UCB/EERC-89/14, October, 1989.

Earthquake Engineering Research Institute, EERI 1989, EERI-NRC-NCEER Congressional Briefing, "Loma

Prieta (Northern California) Earthquake," Senate Russell Office Building, Room 253, November 30, 1989.

Priestley, M. J. N., Seible, P., and Chal, Y. H., 1989, "Collapse of the Cypress Viaduct," Final Report to

Caltrans Office of Structural Design, November 9, 1989.

Uniform Building Code, 1988 Edition, International Conference of Building Officials, 5360 South Workman

Mill Road, Whittier, CA 90601, Section 2312.

5-92
6. PERFORMANCE OF FIRE PROTECTION SYSTEMS

by Harold E. Nelson

6.1 Introduction

In general, private fire protection facilities survived the earthquake without interruption while public fire

protection systems were severely interrupted. Private fire protection systems are limited to a single facility

and mostly within buildings while public fire protection systems are external, community wide, and often

underground. Several affected communities, including San Francisco, were left in a condition where it is

doubtful that they could have halted a serious spreading fire. The absence of a serious spreading fire

(other than the one in the Marina District of San Francisco) is believed to be a combination of factors

including:

o Prompt shut down of electric power (by the utilities).

o Modern safeguards on burners and pilots on water heaters and other gas fired appliances

in service at the time,

o Warm weather (i.e. lack of use of heating equipment).

° Absence of wind (a major fire spread factor if present).

° High moisture content in ground and wild lands (relates to lack of significant wild fire

problems).

° The occurrence of moderate earthquake with a short duration of strong ground shaking

which caused limited damage on buildings and utility structures in the affected

communities.

o Good fortune,

6.2 General Observations

Fire claimed one life, the destruction of a row of about eight apartment buildings in the Marina District,

the loss of a auto repair facility in Berkeley, the loss of a single-family dwelling in Los Gatos, and the

loss of a single-family dwelling in Watsonville. In general, the losses due to fire were minor as compared

to structural damage in this earthquake. This is in contrast to the San Francisco earthquake of 1906

where fire was a major factor. A number of factors including force of the earthquake, wind conditions,

moisture content in the wild lands, and differences in the type and arrangement of heating, lighting, and

cooking facilities.

6-1
It is fortunate that the number of fires were small. Throughout most of the areas struck by the

earthquake, the water supplies necessary to combat a significant fire were disrupted and the

communications necessary to secure extra resources were either disrupted or so overtaxed as to be of

greatly reduced value. In some cases, notably San Francisco and Watsonville, the combination of disrupted

facilities and overloaded radio channels virtually halted emergency communications by the fire department.

Throughout the affected region, except Oakland, the power distribution within the communities was

immediately ceased. This action eliminated electric power as an ignition source and caused all electrically

powered safety valves, (e.g., safety pilots on water heaters, etc.) to close shutting off the pilot and

interrupting the flow of gas at that point. It is believed that this action prevented many ignitions in those

buildings where there was enough movement to break the internal pipes and sever the electrical wiring.

Since the earthquake occurred during daytime, the loss of illumination resulting from the power outage

did not appear to have an impact on the safety of individuals.

Except for Oakland, where underground breakage appears to have been very limited, all of the communities

visited suffered sufficient underground breakage of utilities to disrupt the water mains including a 12 in.

high-pressure line in San Francisco, thereby eliminating the principal source of fire fighting water. Also,

the gas mains were disrupted and gas leaks were prevalent, though the removal of power sources and the

caution exercised by the citizens was effective in preventing ignition of leaking gas. It is likely that the

leakage was distributed widely over the systems, probably concentrated at the points where smaller service

pipes are connected to mains. The gas was leaked in a manner that allowed rapid dispersal in the

atmosphere. In such case, the odorizing agent tends to remain and encouraged continued precautions.

6.3 Meteorological Effects

At the time of the earthquake and through the next several days there was little wind. This is felt to have

been an important factor in the fire which occurred in the Marina District. Had there been a wind, it

is quite possible that fire could have developed into a multi-block conflagration. In addition, rains

immediately prior to the earthquake had resulted in a high moisture in the ground and wild lands. In

areas such as the Santa Cruz mountains, there were ignition sources resulting from downed power lines.

Some minor fires did occur. One was significant but still localized. Had the hills been dry and/or a strong

wind been present, a different result could well have occurred.

6-2
1

6.4 Private Fire Protection Systems

The initial data indicates that private fire protection systems such as sprinklers, standpipes, and alarm

systems for the most part were not interrupted. This is probably due to both the earthquake bracing

requirements included in the fire protection standards governing the installation of these devices in

earthquake zones and the fact that most of the buildings where such equipment is involved did not suffer

extensive structural damage.

6.5 Summary

This earthquake did not place as much stress on the fire-fighting system as a major earthquake would have.

If a major earthquake were to occur, the increased level of destruction of buildings and the resulting

increase of available fuel and ignition sources would probably increase the number of fires. If electric

power is removed as promptly in such an earthquake, it is likely that many of the potential ignition

sources will be interrupted as apparently occurred in this earthquake.

The underground water distribution and gas distribution systems are very vulnerable and generally showed

extensive failure even in this moderate earthquake. It appears that the advances in survival of these

systems has not matched the advances in the survival of structures. It is critical to conduct further

investigation of the impact of the earthquake on the existing systems and the development of better means

to insure continued fire fighting water supply and preventing or minimizing the leakage of gas from the

distribution system is critical.

6-3

7. CONCLUSIONS AND RECOMMENDATIONS

by Richard N. Wright and H. S. Lew

7.1 Conclusions

The Loma Prieta earthquake of October 17, 1989, can provide significant lessons for public policies and

construction practices throughout the United States. While it was the largest earthquake to strike

Northern California since the great San Francisco earthquake of 1906, earthquakes of similar or larger

magnitude are expected to affect 46 of the 50 states, Puerto Rico and the Virgin Islands (Federal

Emergency Management Agency, 1988). Thus, lessons of structural performance, seismology, and geology

which can be learned from the Loma Prieta earthquake, can be used to reduce earthquake hazards

throughout the United States and the rest of the world.

7.1.1 Performance of All Structures

1. The Loma Prieta earthquake caused human losses of 62 lives and 3,757 injuries. But for fortunate

circumstances, human losses would have been much greater. The unusual event of coincidence

with a Bay area World Series baseball game reduced traffic far below normal levels on the 1-880

viaduct which collapsed to cause most of the life losses. Although water supplies for fire fighting

were lost over much of the Bay area, unusually calm winds avoided uncontrolled spreading of fires.

The areas affected by the earthquake are among the best prepared in the United States. Clearly,

earthquakes remain the United States' greatest single-event natural hazard threatening losses of

life and injuries.

2. Direct property losses are estimated to exceed $ 6 billion, but these are modest compared to losses

projected for earthquakes closer to major metropolitan areas, for instance $40 billion for a

magnitude 7.5 earthquake on the eastern Bay area's Hayward fault, (California Department of

Conservation, 1987) or $37 billion for a magnitude 7.6 earthquake in the New Madrid area of the

Central U.S. (FEMA, 1985). Unless actions are taken to reduce earthquake hazards, nationwide,

earthquake property losses can be sufficiently large to seriously impact the U.S. economy (The

Earthquake Project, 1989).

3. Consequent economic losses were apparent in the major disruptions of transportation in the month

that the San Francisco-Oakland Bay Bridge was out of service. Dislocations continue with major

segments of the Bay area highway system still not functioning. The potential for consequent

7-1
economic, national security and societal losses arising from earthquake damages add to the status

of earthquakes as the Nation's greatest single-event natural hazard.

4. Most building and lifeline structures in the epicentral and Bay areas performed well. Californian

efforts to develop, adopt and enforce up to date seismic design and construction practices for

buildings and lifelines deserve great credit for this. However, much structural damage and most

losses of lives and property occurred in the Bay area 60 to 70 miles from the earthquake source.

At such distances, for this sized earthquake, current seismic design and construction practices were

expected to be capable of preventing structural damages. Thus, structural performance in the

Loma Prieta earthquake provides important opportunities for the assessment and improvement of

national seismic design and construction practices.

5. Most structural damages in the Bay area, 60 to 70 miles from the epicentral area, occurred to

structures sited on deep deposits of overburden soils above the bedrock. Such sites are common

in other areas of the Nation that also are subject to earthquake hazards. Also, the intensity of

bedrock ground shaking was much higher in the Bay area than in other directions from the

epicentral area possibly due to directional characteristics of the rupture. Because many records

of strong ground motions are available from the main shock and aftershock measurements, there

important opportunities exist to develop and verify seismological and engineering practices for

prediction of the effects of geological and overburden soil conditions on the intensity of earthquake

ground motions.

6. Structural damages and failures in both the epicentral and Bay areas indicate some vulnerability

for structures having natural periods less than that of their sites. In-depth studies of responses

of damaged and undamaged structures can assess the needs, if any, for revisions of current seismic

standards which allow reduced design forces for structures having short natural periods sited on

deep, soft soil deposits.

7. Ground failures, including landslides and liquefaction, were frequent in both epicentral and Bay

areas. These attest to the potential of ground failures to produce earthquake losses of lives and

property and provide opportunities to assess the needs to improve existing criteria for the stability

of soils in earthquakes.

7-2
7.1.2 Buildings Performance

1. Buildings designed and constructed in accord with modern seismic design and construction practices

generally performed well in both epicentral and Bay areas. This shows the value of adopting and

enforcing modern seismic design and construction practices in building codes and in Federal

construction. However, examples of successful performance in areas of strong shaking and of

failures in areas of moderate shaking provide opportunities to assess needs to improve modern

seismic design and construction practices.

2. Widespread failures of older buildings, such as collapses of unreinforced masonry buildings, failures

of cripple stud walls, and losses of chimneys and parapets, reemphasize the need to assess the

safety of existing buildings and to correct excessive hazards. It is sobering to note that many

modern buildings also have these vulnerabilities in seismically hazardous areas of the United

States which have not yet, or only recently have, adopted and enforced seismic design and construction provisions in their building codes.

3. Many older buildings in the epicentral and Bay areas had been retrofitted to improve seismic

resistance prior to the earthquake. Some performed well; some failed. In-depth studies of the

performance of retrofit measures can reveal opportunities to affirm or to improve standards for

assessment and strengthening of existing buildings that currently are being developed.

4. Many buildings in urban areas were damaged by pounding against adjacent buildings. In-depth

studies of occurrences and non-occurrences of pounding can provide improvements, where needed,

in standards for design of new buildings, and in standards for assessment and strengthening of

existing buildings.

7.1.3 Lifelines Performance

1. The severe losses of lives in the collapse of the 1-880 viaduct, which were fortuitously far less than

the losses would have been under more normal traffic patterns, show the importance of successful

performance of lifelines for life safety.

I

7-3
2. The disruptions of economic activities in the Bay area caused by failures of highway structures, and

in the epicentral area caused by the landslides that closed state Highway 17, show the importance

of successful performance of lifelines for the avoidance of property losses and consequent damages.

3. Water distribution system failures in the Bay area show the vulnerability of U.S. cities in areas

of seismic hazard to fires following earthquakes. Gas distribution system failures in the Bay and

epicentral areas show a significant potential for ignition of fires following earthquakes. These

failures were predominately in areas of unstable soils.

7.2 Recommendations

These recommendations are based on the information gained in investigations to date of the Loma Prieta

earthquake and on the state of the art of earthquake hazards reduction practices.

7.2.1 Performance of All Structures

1. In all seismically hazardous areas of the United States, Federal agencies responsible for the

seismic safety of Federal or Federally assisted or regulated structures, and state and local

governments responsible for regulations for buildings and lifelines should:

Plan and initiate systematic programs to assess the safety of existing buildings and lifelines

and to correct those that are considered to be unduly dangerous. Particular attention

should be given to structures sited on deep soil deposits which can be vulnerable to both

nearby and distant earthquakes. Losses of lives and property and consequent damages

in the Loma Prieta earthquake were due primarily to failures of structures not designed

and constructed in accord with modern practices for seismic safety. To a large extent,

these damages occurred to structures located on deep soil deposits. Those in the Bay area

were far from the earthquake source.

Adopt and enforce modern seismic design and construction practices for new buildings and

lifelines. Generally good performance of modern buildings in the epicentral area of the

Loma Prieta earthquake shows the efficacy of modern design and construction practices.

Tragic losses of 242,000 lives in Tangshan, China in 1976 (China Academy of Building

Research, 1986) and about 30,000 lives in Soviet Armenia in 1988 (Wyllie and Filson,

7-4
1989) show that earthquake losses can be catastrophic in the absence of the adoption and

enforcement of good seismic design and construction practices.

Adopt and enforce modern practices for assessment of the stability of slopes and areas

potentially susceptible to liquefaction to assure that undue existing hazards are identified

and corrected and that new construction is not subject to ground failures in earthquakes.

Significant property and consequent losses from ground failures in the homa Prieta

earthquake, and severe life losses in earlier earthquakes show the importance of prevention

of ground failures affecting buildings and lifelines.

2. In-depth studies should be made of amplifications of ground shaking by overburden soils and

improved practices should be developed for site-specific design criteria for earthquake loadings.

Information from the Loma Prieta earthquake on main and aftershock ground shaking intensities

and on structural performance on deep soil deposits provides a strong data base for such studies.

These data can be combined with similar data obtained from the 1967 Caracas, 1985 Mexico, and

1988 Spitak earthquakes.

3. In-depth studies should be made of the effects of geologic structures and earthquake mechanisms

on the directionality of the propagation of strong shaking from earthquake fault areas. The

marked irregularity of the intensity contours for the Loma Prieta earthquake show large potential

for savings in costs of earthquake resistance and for reductions in earthquake damages when

intensity distributions can be predicted more accurately for use in siting and design of structures.

4. Methods should be developed for measurement of the properties of existing structures that will

allow reliable pre-earthquake assessment of vulnerability and post-earthquake assessment of

damages. Experiences in the Loma Prieta earthquake reemphasize both the need for these

assessments and the inadequacies of currently available technologies.

7.2.2 Buildings Performance

1. Successful and unsuccessful performances of modern buildings in the Loma Prieta earthquake

should be studied in depth to formulate improvements for seismic design and construction

practices. Recommendations from development of the National Earthquake Hazards Reduction

Program (NEHRP) Provisions, (FEMA 1988), emphasize the need to improve standards for the
energy absorbing capacities of various structural systems. The large variety of quantitative

information on building performance in the Loma Prieta earthquake provides data for assessing

and confirming analytical models for nonlinear structural performance in earthquakes that in turn

can be used in development of improved design criteria.

2. Successful and unsuccessful performances of buildings that had been retrofitted prior to the Loma

Prieta earthquake should be studied in depth to confirm the validity of, or to improve, standards

for the assessment and strengthening of existing buildings that presently are being developed in

the NEHRP. The large variety of data on performance of retrofit buildings available from the

Loma Prieta earthquake provides a significant opportunity to evaluate and improve practices for

the assessment and strengthening of existing buildings.

3. Techniques should be developed for reduction of pounding damages for new and existing buildings.

Substantial pounding damages were observed in the Loma Prieta and in the Mexico City

earthquake of 1985. Data are newly available for assessment of the adequacy of current criteria

for new buildings design. Research can devise and verify techniques for mitigating potential

pounding damages between existing buildings.

7.2.3 Lifelines Performance

1. Analytical and laboratory studies of highway structures' supports should be conducted to improve

design, assessment and retrofit practices. Severe collapses and damages in the Loma Prieta

earthquake and the fact that most of the U.S. highway system has not been designed for

earthquake resistance show the need for these studies.

2. Analytical, laboratory and field studies of the performance of gas and water pipelines should be

conducted to develop improved design, assessment and retrofit practices. Extensive failures and

consequent fire hazards occurred in areas of unstable soils in the Loma Prieta earthquake. Similar

potentials exist in most seismically hazardous areas of the United States.

3. Nationally applicable design and construction provisions for new lifelines, and assessment and

strengthening provisions for existing lifelines should be developed in the NEHRP. The Loma

Prieta earthquake shows the importance of successful lifelines performance for the mitigation of

life, property and consequent losses due to earthquakes. In contrast to buildings, and except for

7-6
highway structures, no nationally applicable design and construction practices are available for new

and existing lifelines.

7.3 References

Federal Emergency Management Agency, FEMA 1985, "An Assessment of Damage and Casualties for Six

Cities in the United States Due to Earthquakes in the New Madrid Seismic Zone," FEMA, Central

United States Earthquake Preparedness Project.

California Department of Conservation, 1987, "Earthquake Planning Scenario for a Magnitude 7.5

Earthquake on the Hayward Fault in the San Francisco Bay Area," Special Publication 78.

China Academy of Building Research, 1986, "The Mammoth Tangshan Earthquake of 1976," China

Academic Publishers.

The Earthquake Project, 1989, "News."

Federal Emergency Management Agency, 1988, "NEHRP Recommended Provisions for the Development

of Seismic Regulations for New Buildings," FEMA 95.

Wyllie, L. A, and J. R., Filson, 1989, "Armenia Earthquake Reconnaissance Report,"

Earthquake Spectra, August 1989.

7-7

4. TITLE AND SUBTITLE

Performance of Structures During the Loma Prieta Earthquake of October 17, 1989

5. AUTHOR(S)

H. S. Lew, Editor

6. PERFORMINQ ORQANIZATION (IF JOINT OR OTHER THAN NIST, SEE INSTRUCTIONS)

U.S. DEPARTMENT OF COMMERCE

NATIONAL INSTITUTE OF STANDARDS AND TECHNOLOGY

QAITHERSBURQ, MD 20899

7. CONTRACT/GRANT NUMBER

a. TYPE OF REPORT AND PERIOD COVERED

Final

NIST-114A

(REV. 3-89)

U.S. DEPARTMENT OF COMMERCE

NATIONAL INSTITUTE OF STANDARDS AND TECHNOLOGY

BIBLIOGRAPHIC DATASHEET

1. PUBUCATION OR REPORT NUMBER

NIST/SP-778

2. PERFORMINQ ORGANIZATION REPORT NUMBER

3. PUBUCATION DATE

January 1990

9. SPONSORING ORGANIZATION NAME AND COMPLETE ADDRESS (STREET, CITY, STATE, ZIP) Dept of the Interior

U.S. Geological Survey

582 National Center

Reston, VA 22092

10. SUPPLEMENTARY NOTES

DOCUMENT DESCRIBES A COMPUTER PROGRAM; SF-185, FIPS SOFTWARE SUMMARY, IS ATTACHED.

11. ABSTRACT (A 200-WORD OR LESS FACTUAL SUMMARY OF MOST SIGNIFICANT INFORMATION. IF DOCUMENT INCLUDES A SIGNIFICANT BIBUOGRAPHY OR

UTERATURE SURVEY, MENTION IT HERE.)

Immediately following the magnitude 7.1 Loma Prieta earthquake of October 17, 1989/ a team

representing the Interagency Committee on Seismic Safety in Construction surveyed the

damage to buildings, utilities and transportation structures. This report is based

primarily on the data gathered during the site survey and the results of preliminary analyses

of structural failures. Most structures designed in accordance with modem codes and

standards performed well without serious structural damage. However, there were many

concrete and masonry buildings and highway structures in the San Francisco Bay area which

were not designed according to modem seismic design codes and which did not perform well.

The majority of damaged structures had not been strengthened to increase earthquake

resistance. Except for two deaths from landslides and one from fire, the remainder of

the 62 deaths from the earthquake was due to partial or total collapse of older structures.

This investigation provided the basis for recommendations to improve design and construction

practices for buildings and lifeline structures and to mitigate damage to existing structures

in future earthquakes.

12. KEY WORDS (6 TO 12 ENTRIES; ALPHABETICAL ORDER; CAPITAUZE ONLY PROPER NAMES; AND SEPARATE KEY WORDS BY SEMICOLONS)

Bridges, buildings; codes; damage; earthquakes; fire safety; geology; highways;

housing; lifelines; seismic; specifications; structural engineering; viaducts.

13. AVAILABIUTY ^

UNUMITED

FOR OFFICIAL DISTRIBUTION. DO NOT RELEASE TO NATIONAL TECHNICAL INFORMATION SERVICE (NTIS).

1.

ORDER FROM SUPERINTENDENT OF DOCUMENTS, U.S. GOVERNMENT PRINTING OFFICE,

WASHINGTON, DC 20402. ORDER FROM NATIONAL TECHNICAL INFORMATION SERVICE (NTIS), SPRINGFIELD, VA 22161.

14. NUMBER OF PRINTED PAGES

201

15. PRICE

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