Wind and Seismic Effects

Proceedings of the 34th

Joint Panel Meeting

NIST Special Publication 987

Stephen A. Cauffman, Editor

US7

National Institute of Standards and Technology • Technology Administration • U.S. Department of Commerce
rhe National Institute of Standards and Technology was established in 1988 by Congress to "assist industry in

the development of technology . . . needed to improve product quality, to modernize manufacturing processes,

to ensure product reliability . . . and to facilitate rapid commercialization ... of products based on new scientific

discoveries."

NIST, originally founded as the National Bureau of Standards in 1901, works to strengthen U.S. industry's

competitiveness; advance science and engineering; and improve public health, safety, and the environment. One

of the agency's basic functions is to develop, maintain, and retain custody of the national standards of

measurement, and provide the means and methods for comparing standards used in science, engineering,

manufacturing, commerce, industry, and education with the standards adopted or recognized by the Federal

Government.

As an agency of the U.S. Commerce Department's Technology Administration, NIST conducts basic and

applied research in the physical sciences and engineering, and develops measurement techniques, test methods, standards, and related services. The Institute does generic and precompetitive work on new and

advanced technologies. NIST's research facilities are located at Gaithersburg, MD 20899, and at Boulder, CO 80303.

Major technical operating units and their principal activities are listed below. For more information contact the

Publications and Program Inquiries Desk, 301-975-3058.

Office of the Director

• National Quality Program

• International and Academic Affairs

Technology Services

• Standards Services

• Technology Partnerships

• Measurement Services

• Information Services

Advanced Technology Program

• Economic Assessment

• Information Technology and Applications

• Chemistry and Life Sciences

• Electronics and Photonics Technology

Manufacturing Extension Partnership

Program

• Regional Programs

• National Programs

• Program Development

Electronics and Electrical Engineering

Laboratory

• Microelectronics

• Law Enforcement Standards

• Electricity

• Semiconductor Electronics

• Radio-Frequency Technology'

• Electromagnetic Technology 1

• Optoelectronics

1

• Magnetic Technology 1

Materials Science and Engineering

Laboratory

• Intelligent Processing of Materials

• Ceramics

• Materials Reliability'

• Polymers

• Metallurgy

• NIST Center for Neutron Research

Chemical Science and Technology

Laboratory

• Biotechnology

• Physical and Chemical Properties

2

• Analytical Chemistry

• Process Measurements

• Surface and Microanalysis Science

Physics Laboratory

• Electron and Optical Physics

• Atomic Physics

• Optical Technology

• Ionizing Radiation

• Time and Frequency

1

• Quantum Physics'

Manufacturing Engineering

Laboratory

• Precision Engineering

• Manufacturing Metrology

• Intelligent Systems

• Fabrication Technology

• Manufacturing Systems Integration

Building and Fire Research

Laboratory

• Applied Economics

• Structures

• Building Materials

• Building Environment

• Fire Research

Information Technology Laboratory

• Mathematical and Computational Sciences

2

• Advanced Network Technologies

• Computer Security

• Information Access

• Convergent Information Systems

• Information Services and Computing

• Software Diagnostics and Conformance Testing

• Statistical Engineering

'At Boulder, CO 80303.

2 Some elements at Boulder, CO.
NIST Special Publication 987

Wind and Seismic Effects

Proceedings of the 34th Joint Panel Meeting

Stephen A. Cauffman, Editor

Materials and Construction Research Division,

Building and Fire Research Laboratory,

National Institute of Standards and Technology,

Gaithersburg, MD 20899-8611

September 2002

**tes o* *

U.S. Department of Commerce

Donald L. Evans, Secretary

Technology Administration

Phillip J. Bond, Under Secretary for Technology

National Institute of Standards and Technology

Arden L. Bement, Jr., Director
Certain commercial entities, equipment, or materials may be identified in this document in order to describe an experimental procedure or concept adequately. Such

identification is not intended to imply recommendation or endorsement by the

National Institute of Standards and Technology, nor is it intended to imply that the

entities, materials, or equipment are necessarily the best available for the purpose.

National Institute of Standards and Technology Special Publication 987

Natl. Inst. Stand. Technol. Spec. Publ. 987, 382 pages (September 2002)

CODEN: NSPUE2

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 2002

For sale by the Superintendent of Documents, U.S. Government Printing Office

Internet: bookstore.gpo.gov —Phone: (202) 5 1 2- 1 800 —Fax: (202) 5 1 2-2250

Mail Stop SSOP, Washington, DC 20402-0001
PREFACE

This publication contains the Proceedings of the 34 Joint Meeting of the U.S.-Japan Panel on Wind and

Seismic Effects. The technical meetings were held at the National Institute of Standards and Technology

in Gaithersburg, Maryland during 13-15 May 2002 followed by three-day technical site visits. Thirty

papers were written, 15 by each side. Twenty-eight papers were presented orally; 15 by the U.S. -side and

13 by the Japan-side. The papers were organized into five Sessions: Geotechnical Engineering and Ground

Motion; Next-Generation Building and Infrastructure Systems; Wind; Dams; and Transportation Systems

and three Technical Themes: Terrorism ~ Collapse of the WTC and Pentagon Wedge A; Disaster

Information; and Management of Infrastructure Systems.

BACKGROUND

Responding to the need for improved engineering and scientific practices through exchange of technical

data and information, research personnel, and research equipment, the United States and Japan in 1961

created the U.S.-Japan Cooperative Science Program. Three collateral programs comprise the Cooperative

Science Program. The U.S.-Japan Cooperative Program in Natural Resources (UJNR), one of the three,

was created in January 1964. The objective of UJNR is to exchange information on research results and

exchange scientists and engineers in the area of natural resources for the benefit of both countries. UJNR is composed of 16 Panels each responsible for specific technical subjects.

The Panel on Wind and Seismic Effects was established in 1969. Eighteen U.S. and 14 Japanese

organizations participate with representatives of private sector organizations to develop and exchange

technologies aimed at reducing damages from high winds, earthquakes, storm surge, and tsunamis. This

work is produced through collaboration between U.S. and Japanese member researchers working in five

Task Committees. Each committee focuses on specific technical issues, e.g., earthquake strong motion

data. The Panel provides the vehicle to exchange technical data and information on design and

construction of civil engineering lifelines, buildings, and water front structures, and to exchange high wind

and seismic measurement records. Annual meetings alternate between Japan and the United States (odd

numbered years in Japan; even numbered years in the United States). These one-week technical meetings

including technical site visits that follow the meetings provide the forum to discuss ongoing research and

research results. The National Institute of Standards and Technology (NIST) provides the U.S. -side chair

and secretariat. The Public Works Research Institute (PWRI), Japan, provides the Japan-side chair and

secretariat.

Task Committee meetings, exchanges of data and information through technical presentations at annual

Panel meetings, exchanges of guest researchers, visits to respective research laboratories and informal

interactions between Panel meetings, joint workshops and seminars, and joint cooperative research

programs all contribute to the development and effective delivery of knowledge that has influenced design

and construction practices in both countries. Guest research exchanges have advanced the state of

technology in areas of steel, concrete, and masonry structures under seismic forces; developed techniques

to analyze risks from liquefaction; modeled water seepage in dam foundations; performed comparative

analyses of seismic design of U.S. and Japanese bridges.

Direct communication between counterpart country organizations is the cornerstone of the Panel. Effective

information exchanges and exchanges of personnel and equipment have strengthened domestic programs

of both countries. There are opportunities for experts in various technical fields to get to know their

foreign counterparts, conduct informal exchanges, bring their respective views to the frontiers of

• • • 111
knowledge, and advance knowledge of their specialties.

The Panel's activities resulted in improved building and bridge standards and codes and design and

construction practices in hydraulic structures in both countries, for example:

• created and exchanged digitized earthquake records used as the basis of design and research in

Japan and the United States;

• transferred earthquake engineering information and strong-motion measurement techniques for use by seismically active countries, e.g., Australia, Canada, Italy, Mexico, Peru, Taiwan, Turkey, and

North Africa;

• produced data that advanced retrofit techniques for bridge structures;

• developed field test data for use in aerodynamic retrofit of bridge structures;

• produced full-scale test data that advanced seismic design standards for buildings;

• advanced technology for repairing and strengthening reinforced concrete, steel, and masonry

structures;

• improved in-situ measurement methods for soil liquefaction and stability under seismic loads;

• created a database comparing Japanese and U.S. standard penetration tests to improve prediction

of soil liquefaction;

• created a database on storm surge and tsunamis and verified mathematical models of tsunami and

storm surge warning systems;

• established a library resource of current research on wind and earthquake engineering and on

storm surge and tsunamis;

• published proceedings of Panel meetings, Task Committee Workshops, and special publications

such as List of Panel Publications and translated from Japanese to English a two-volume series on

earthquake resistant construction using base isolation systems;

• gained better knowledge of both countries' research, design, and construction capabilities from indepth visits to the host country's laboratories and building and public work projects. Results of

such visits contribute to creation of new Task Committees, agendas for Joint Panel meetings and

task committee workshops, special visits of U.S.-Japan researchers, and joint collaborative

research.

HIGHLIGHTS OF THE TECHNICAL SITE VISITS

Four technical sites visits occurred during 16-18 May 2002 at three locations.

New York City

1. World Trade Center

On 16 May 2002, the Panel delegation visited the engineering firm of LZA Technologies, a division of the

Thornton Tomasetti Group, New York, NY to discuss the structural engineers response to the World Trade

Center emergency, reconstruction plans, recommendations for risk management, and other lessons learned

from the disaster. Mr. Donald Freeman, Director of Preservation and Senior Associate welcomed the

delegation and discussed the work that was performed to clean up the wreckage from the collapse of the

WTC towers One and Two, and other related damaged structures. Key points from the briefing include:

• Subsequent to the September 1

1

th attacks, 450 buildings were investigated for safety concerns,

• The towers were constructed on reclaimed land from a 200-year-old landfill, mostly garbage. This

area includes lower Manhattan Island West of Green Street.

i

i

i
• There were three subways crossing the site, of great concern for flooding of these tunnels, which

would have shut down most subway operations in New York City.

• The core columns were 1 m by 2 m by 15 cm thick. They were not designed for lateral loads or for

seismic activities.

• All of the structural steel in the buildings was fire proofed with spray on material. The WTC

Buildings One and Two had asbestos spray on fireproofing up to floor 39 and non-asbestos

fireproofing beyond the 39th floor.

• At the time of the building design, there were no seismic concerns; the outer tubes were designed

to resist all lateral loads.

• Initial assessment from the debris pile shortly after the collapse revealed no fire proofing on any of

the metal. This may be due to the aircraft impact, extremely hot temperatures from burning

aviation fuels, and impact after collapsing; but the cause is uncertain.

• To a large extent the framing was pushed outward away from the direction of each wall.

• Several sections of the outer walls remained intact and survived the collapse.

• Failure to other buildings in the area and damage was caused by falling debris and fire.

• The office responsible for emergency management and actions during a crisis such as this, the Port

Authority of New York and New Jersey, recent former owners of the WTC and designer, and other

organizations had their offices located in the WTC. The collapse severely hampered their ability

to function and caused a long delay in the commencement of rescue operations.

• LZA was contracted by NY City to manage all structural engineering components of the clean up

operations.

• The US Occupational Safety and Health Administration and the Environmental Protection Agency

provided a perimeter for respiratory control due to all the materials and chemicals present in the

debris pile. This perimeter was established two months after the collapse.

• Four engineering/construction firms were sub-contracted to perform the actual clean up activities

under the general control of LZA. The work was subdivided between the four companies. BOVIS

and AMEC shared the responsibility for towers one and two, Turner had responsibility for Tower

7, and Tully Construction Company the responsibility for the rest of the buildings in the 6.4 ha (16

ac) site. SEAoNY (Structural Engineers Association of New York) Teams were operating as subcontractors to LZA Technologies.

• LZA Technologies staff performed structural assessments of 450 buildings; identified unstable

structures; developed demolition and stabilization procedures when taking down collapsed and

damaged buildings; identified safe locations for placing heavy equipment; developed and

coordinated survey monitoring programs; inspected over 400 buildings; performed detailed

investigations of damaged structures.

• Approximately 200 engineers working in 12-hour shifts inspected buildings and investigated the

site during the first week; many were volunteers.

• The buildings that were damaged and removed from the WTC site are 1 , 2, 3, 4, 5, 6, and 7. Building 7 is the one that burned uncontrollably for several hours.

• There is one building, the Bankers Trust Building, which may not be salvageable.

• There is one very old building, 90 West Building, that did extremely well from fire damage due to

the presence of terra-cotta protection.

• The slurry walls were unbraced and very dangerous. The wall at Liberty Street did experience

several centimeters of movement. All of the walls were braced and are stable.

Following the presentation at LZA Technologies, the group was taken to the WTC site to view the ongoing

clean up activities. The tour began in the American Express Building, which is immediately adjacent to

V
the site. From the 26 floor, an excellent view was afforded of the site and all activities. After this, a walking tour of the site was provided, essentially along the west slurry wall of the 160 m (510 ft) by 30 m

(98 ft) by 27 m (98 ft) deep pit where more than 764,555 m

3 (1 million yd3 )of rubble was removed.

Explanations of the ongoing activities were provided.

San-Francisco

2. San Francisco-Oakland Bay Bridge and Golden Gate Bridge

http ://www. mtc.ca. gov/proj ects/bay bridge/bbmain .htm

The delegation was hosted by Mr. Robert Tanaka, California Department of Transportation (CALTRANS)

who provided the delegation a briefing on the East Span Replacement of the Oakland-San Francisco Bay

Bridge, followed by site visits to the San Raphael Bridge and the Golden Gate Bridge. CALTRANS

maintains, operates, and administers seven State-owned toll bridges on San Francisco Bay and one bridge

each in Los Angeles and San Diego.

The West Span Crossing of San Francisco-Oakland Bay Bridge is part of Interstate 80 connecting the city

of San Francisco and Yerba Buena Island. It is 2,870 m (9,260 ft) long and consists of two suspension

bridges. This structure was opened to traffic in 1936 and has an average daily traffic volume of 270,000

vehicles. It is currently undergoing substructure and superstructure seismic retrofitting. The estimate for

seismic retrofit is $250 million. Work began in July 1999; it is scheduled to be completed in April 2003.

Richard Heringer, CALTRANS summarized the work for stiffening the west side of the San FranciscoOakland Bay Bridge superstructure by replacing all bearings with unidirectional isolator bearings that

permitted sliding in the longitudinal direction. Significant retrofit was performed at the base of the towers

and improved the connection of tower to pier. The superstructure retrofit included installing perforated

plates and replacing the 500,000 rivets with round head bolts. Installation of 96 dampers is a major

component of the retrofit.

Following the 1989 Loma Prieta earthquake that collapsed a section of the east span of the San FranciscoOakland Bay Bridge, CALTRANS started work to replace the East Span to meet current seismic standards.

CALTRANS undertook an extensive design and public review process. Running from Yerba Buena Island

to Oakland, the new structure will feature a self-anchored, single-tower suspension span across the

shipping channel transitioning to a skyway. The current bridge is double-decked; the replacement will

feature side-by-side decks. A 5 m (16 ft) wide bicycle and pedestrian path will run along the eastbound

deck of the 3.2 km (2 mi) long structure. Its tower is 162 m (525 ft) above the water. This new structure

will be on Interstate 80 and connect Yerba Buena Island and the city of Oakland. It will consist of a self- anchored suspension bridge section and post-tensioned concrete box-girder skyway section. This structure

will accommodate an average daily traffic volume of 270,000 vehicles. The present estimate for the

construction of this structure is $2.6 billion. Work began in February 2002 and is scheduled for completion

in 2007.

The delegation visited the San Raphael Bridge $400 million dollar seismic retrofit project where its substructure was extensively modified. This structure is on Interstate 580 and connects the cities of

Richmond and San Rafael. It is 5.5 miles long and consists of two steel cantilever bridge sections and

many steel truss sections. This structure was opened to traffic in 1956 and has an average daily traffic

volume of 60,000 vehicles. It is currently undergoing substructure and superstructure seismic retrofitting.

The present estimate for seismic retrofit contract is $484 million. Work began in August 2000 and is scheduled to be completed in December 2004.

vi
Lastly, the delegation visited the Golden Gate Bridge where they discussed its seismic retrofit project. Mr.

Dennis Mulligan, chief engineer of the seismic retrofit project welcomed the delegation. Following the

1989 Loma Prieta earthquake the Golden Gate Authority began a study to retrofit the bridge. It was divided

into three phases: north viaduct; south structure; and the suspension bridge itself. Dampers were installed

at the towers and bridge truss and at the arch pylon/bridge trusses. The towers were stiffened, the

saddle/cable connection was strengthened, and isolators were installed on the viaduct. The cables were

inspected in 1991 and 2000 and were assessed to be in good condition.

Golden Gate Bridge

The Golden Gate Bridge is internationally known as one of the greatest suspension bridges ever built.

Since opening in 1937, it has served as a critical link in California's highway system and as an engine of

economic vitality, supporting the region's industry, commerce and tourism. The Golden Gate Bridge now

carries 42 million vehicles a year and is visited by more than 10 million people annually. The Golden Gate

Bridge remains vulnerable to earthquakes and could fail during an earthquake of magnitude 7.0 or greater on the nearby San Andreas or Hayward faults. The U.S. Geological Survey reports a 70 % probability of a magnitude 6.7 or greater earthquake striking the Bay Area prior to 2030. Under the leadership of Bay Area

Congressional members the Golden Gate Bridge, Highway and Transportation District (District), the

public agency that built, operates and maintains the Bridge, is seeking an additional $119 million in federal

allocations. Once completed, the $388 million project will protect the Golden Gate Bridge from nearby

earthquakes up to magnitude 8.3.

The Golden Gate Bridge seismic retrofit is separated into three construction phases. Phase 1 retrofitted the

North Viaduct and was completed in December 2001. The second phase began in summer 2001 and will

retrofit the South Viaduct, Anchorage, Fort Point arch, and South Pylons, while the third phase, once

funded, will retrofit the North Anchorage, the Main Span and Main Towers. The retrofit design maintains

the Golden Gate Bridge's aesthetic and historic integrity, while strengthening the Bridge to withstand a maximum credible earthquake.

CALTRANS is constructing a new Carquinez Bridge that the delegation did not visit. This new structure

will be on Interstate 80 near Vallejo, CA. It will be a suspension bridge 1.3 km (0.8 mi) long. This

structure will have an average daily traffic volume of 1 16,000 vehicles. The present estimate for the

construction of this structure is $188 million. Work began in January, 2000 and is scheduled to be

completed in December, 2003.

Reno, Nevada

3. University of Nevada-Reno

The delegation visited the University of Nevada-Reno (UNR) campus in the morning of May 18, 2002. Dr.

Ian Buckle, Director, Center for Civil Engineering Earthquake Research, University of Nevada-Reno, NV

http://bric.ce.unr.edu/cceer.htm hosted the delegation. Following a welcome by Dr. Maragakis, Dean of the

Civil and Environmental Engineering Department, the delegation discussed recent research projects in

earthquake engineering, participated in short presentations on recent research, and observed current

upgrades to their new Multi-Shake Table Facility.

The delegation first visited the Mackay Building, which currently houses the engineering library and the

museum of mining and geology. The Mackay building is an un-reinforced masonry building constructed in

Vll
1908. It is the second oldest building on campus. It is on the National Register of Historic Buildings.

Since it is on the National Register, options for seismic retrofit were limited; thus base isolation was

selected. Completed in 1993, the Mackay Building retrofit uses a combination of high-damping rubber

isolators and friction dampers. The building rests on sixty high-damping rubber isolators at the edges of the

building. Forty friction dampers were installed to support the floor and to reduce the floor span since the

building live loads are relatively high. The stairs at the front of the building are now cantilevered from the

building frame. The building is instrumented to record its response during seismic events.

UNR received a Network for Earthquake Engineering Simulation (NEES) award from the National

Science Foundation (NSF) to invest in equipment to upgrade experimental earthquake simulators. $100

million was provided to 10 universities. UNR received $7 million for upgrades: $4.4 million from NSF;

$1.6 million from the Department of Housing and Urban Development; and $1 million from the

Department of Energy.

The delegation participated in presentations on some of the research being conducted at UNR. Hisham

Nada presented research that is being conducted for CALTRANS to improve the seismic performance of

flared bridge columns. During a seismic event, the plastic hinge shifts down and failure of the column

occurs just below the flare. CALTRANS has issued new guidance for construction of flared columns that

creates a gap between the beams and columns. UNR is investigating the efficiency of the new detail, the

effect of lowering the confinement ratio at the top of the column, and the effect of column height on the

shear level. Experimental work in support of this research is being conducted on 1/5-scale specimens.

Lyle Carden presented his research on improving the seismic performance of steel girder bridges through

the application of ductile substructures and seismic isolation. The application of ductile substructure and

seismic isolation is to assure that the bridge superstructure remains elastic during a seismic event. The

objectives of this work are to investigate the load path and the behavior of components, demonstrate

improved performance through the use of ductile members in the substructure and seismic isolation.

Experimental work has been performed on 2/5 -scale, two girder specimens, with static loads applied at the

deck level. Dynamic testing will be performed on the shake facility at UNR.

Lyle Carden presented a second paper on research performed for CALTRANS related to the replacement

of the East Span of the San Francisco-Oakland Bay Bridge. UNR is evaluating the shear link that will be

used in the tower for self-supporting suspension span. UNR has conducted two tests on the shear link to

date. UNR is also evaluating high performance steel A709 HPS70W for energy dissipation under

earthquake loading for the East Span replacement program.

Fabio Sanchez-Camargo made the final presentation on his research into the seismic performance of bridge

restrainers at intermediate hinges. In multi-span bridges, out-of-phase movement of adjacent frames can

lead to excessive displacement. This research, focused on concrete bridges, will result in a model of typical

hinge details for evaluation of restrainers under seismic loadings.

The visit to UNR concluded with a tour of the structures laboratory. The laboratory was constructed in

1992. The laboratory includes a dual shake table facility. The shake tables became operational in 1998.

The existing shake table facility is being upgraded to add a third shake table. Additional upgrades are being made to give all three-shake tables biaxial loading capability. The upgrades are being funded

through the National Science Foundation's Network for Earthquake Engineering Simulation (NEES)

program.

• • • Vlll
4. DIS Systems INC., Sparks, Nevada, www.dis-inc.com.

Mr. Konrad Erikesen, Vice President, Engineering and Marketing, Dr. Amarnath Kasalanati, and Mr.

Danny Sommers hosted the delegation. Since DIS' creation in 1982, as a manufacture of base isolators and

energy dissipators, the organization supplied isolators for more than 100-bridge projects and 80-buildings

worldwide. About 50 % of these buildings are in Japan. Example projects include Oakland City Hall, San

Francisco City Hall, several hospitals and emergency centers, the Asian Art Museum in San Francisco;

several CALTRANS bridges and the Golden Gate Bridge, and the JFK Light Rail System in NY.

The delegation toured DIS' manufacturing plant and discussed its seismic isolator testing capability, with

1.8 x 10

6 kg (2,000 tons) of compression capacity and shear capacity of 4.1 x 10

5 kg (450 tons) with ± 70

cm stroke. DIS maintains an extensive database of tested isolator properties. These isolators absorb up to 40 MPa load and 400 percent shear strain. DIS' largest isolators are 1.5 m diameter, 0.75 m tall, weigh

2.7x1

3 kg (3 tons), and carry 300 tons. The vertical load carrying capacity of these isolators is about 100

kg/cm2 or about 8,274 kPa (1200 psi) axial stress. These elastomeric isolators consist of a central lead core, energy dissipator, that is surrounded by horizontal layers of rubber and steel reinforcing plates capped with

steel load plates top and bottom that are anchored to the column above and the floor below.

DIS monitors many of their installed systems of the past 10 years. They perform 12 year and 6 year aging

tests on their isolators. We were told the isolator stiffness has increased by 2% to 5%.

DIS performs quality control inspection for each isolator throughout its manufacturing process. Individual

inspection records are kept for each inspection. The rejection rate was reported to be about 1 %. DIS has

not registered for ISO 9000 compliance since accreditation is not required in the US market and at this

point in their manufacturing, DIS management does not see a need to be ISO accredited. One of their

largest customers, Japan has approved DIS' isolators through the Building Council of Japan. When the

company increases in size, the Vice President for Engineering and Manufacturing said DIS will apply for

ISO 9000 registration.

DIS is about a $10 million business employing 40 staff; 30 are on the shop floor. Their Headquarters is in

Lafayette, CA. DIS' manufacturing factory is in Sparks, NV where they produce about 1000 isolators

annually.

Disclaimer

Certain trade names and company products are mentioned in the text or identified in an illustration in order

to adequately indicate a manufactured product or specify an experimental procedure or equipment used. In

no case does such identification imply recommendation or endorsement by the National Institute of

Standards and Technology, nor does it imply that the products are necessarily the best available for the

purpose.

ix

ABSTRACT

This publication is the Proceedings of the 34 Joint Meeting of the US-Japan Panel on

Wind and Seismic Effects. The meeting was held at the National Institute of Standards

and Technology, Gaithersburg, Maryland during 13-15 May 2002. The Proceedings

include the technical program, list of members, Panel Resolutions, Task Committee

Reports, and the 30 technical papers authored for this joint meeting.

The manuscripts were presented within five Sessions: Geotechnical Engineering and

Ground Motion; Next-Generation Building and Infrastructure Systems; Wind; Dams; and

Transportation Systems and three Technical Themes: Terrorism — Collapse of the WTC

and Pentagon Wedge A; Disaster Information; and Management of Infrastructure

Systems.

KEYWORDS: Bridges; building performance; disaster information systems; disaster

reduction; earthquakes; geotechnical engineering; ground failures; ground motion;

infrastructure; liquefaction; masonry; measurements; risk assessment; seismic; standards;

structural engineering; and wind loads.

xi

CONTENTS

PREFACE iii

ABSTRACT xi

AGENDA OF JOINT MEETING xvii

LIST OF PANEL MEMBERS

US-Side Panel Members xxv

Japan-side Panel Members xxxi

RESOLUTIONS xlix

MANUSCRIPTS

Technical Session 1 - GEOTECHNICAL ENGINEERING AND GROUND MOTION

Seismic Performance of Urban, Reclaimed, and Port Areas - Full Scale Experiment

Using Blast Technique 5

Takahiro SUGANO and Eiji KOHAMA

A Review ofEmpirical Evidencefor Site Coefficients in Building-Code Provisions 21

Roger BORCHERDT

A Practical Procedure for Formulating Level 2 Earthquake Motions Based on Scenario

Earthquakes 29

Shojiro KATAOKA, Jun MURAKOSHI, and Keiichi TAMURA

Identification ofSite Frequenciesfrom Building Response Records 37

MehmetCELEBI

Evaluation ofEarth Pressure Acting on Embedded Pile-cap Based on Liquefaction Test

Using Large-scale Shaking Table 55

Shuji TAMURA, Chikahiro MINOWA, Kohji TOKIMATSU, Shunji FUJII, Kaeko

YAHATA, and Tomio TSUCHIYA

Technical Session 2 » NEXT-GENERATION BUILDING AND INFRASTRUCTURE

SYSTEMS

Project on 3-D Full Scale Earthquake Testing Facility 67

Keiichi OHTANI, Nobuyuki OGAWA, Tsuneo KATAYAMA, and Heki SHIBATA

Smart Damping Technologies for Dynamic Hazard Mitigation 75

B. F. Spencer Jr.

Research and Development Programs on Timber Structures in Japan 83

Hiroshi ISODA, Hisashi OKADA, Naohito KAWAI, and Nobuyoshi YAMAGUCHI

• • • Xlll
Modeling Regional Economic Resiliency to Earthquakes: A Computable General Equilibrium

Analysis ofLifeline Disruptions 91

Adam ROSE and Shu-Yi LIAO

Technical Session 3 - WIND ENGINEERING

Aerodynamic Stability ofa Super Long-Span Bridge with Slotted Box Girder 107

Hiroshi SATO, Nobuyuki HIRAHARA, Koichiro FUMOTO, Shigeru HIRANO,

and Shigeki KUSUHARA

Wind Effects on Long Span Cable Stayed Bridges: Assessment and Validation 121

Nicholas JONES and Ender OZKAN

Strategy on Wind Engineering Research in BRI and NILIM 133

Hisashi OKADA, Yasuo OKUDA, Hitomitsu KIKITSU, and Masamiki OHASHI

Next Generation Wind Tunnelsfor Simulation ofStraight-Line, Thunderstorm- and TornadoLike Winds 141

Partha P. SARKAR and Fred L. HAAN, Jr.

Technical Session 4 - DAMS

Seismic Performance Evaluation ofIntake Towers 159

Richard C. DOVE and Enrique E. MATHEU

Present State ofMeasurement ofEarthquake Motion and Nation-Wide Networking of

Seismographs at Dam Sites in Japan 167

Tadahiko SAKAMOTO, Shuji TAKASU, Hitoshi YOSHIDA, Yoshikazu

YAMAGUCHI, Takashi SASAKI, and Kenji INAGAKI

Cyclically-Induced Pore Pressure at High Confining Stress 181

Michael K. SHARP and R. Scott STEEDMAN

Technical Session 5 - TRANSPORTATION SYSTEMS

An Evaluation of the Force Reduction Factor in the Force-Based Seismic Design 201

Gakuho WATANABE and Kazuhiko KAWASHIMA

Bolu Viaduct-1 Subjected to Near-Fault Ground Motion 219

Sunwoo PARK, Hamid GHASEMI, Jerry SHEN, and Phillip YEN

2002 Seismic Design Specifications for Highway Bridges 231

Shigeki UNJOH, Sho-ichi NAKATANI, Kei-ichi TAMURA, Jiro FUKUI, and Junichi HOSHIKUMA

Development ofNational Guidelinesfor Seismic Performance Testing 241

W. Phillip YEN, Jerry J. SHEN, and John O'FALLON

xiv
Special Theme 1 - TERRORISM: COLLAPSE of the WTC and PENTAGON WEDGE A

Terrorist Threats Against the Pentagon 249

Robert L. HALL, James T. BAYLOT, Robert J. DINAN, Richard DOVE, John HAYES,

W. Matthew HOSSLEY, James L. O'DANIEL, Thomas R. SLAWSON, and Stanley C.

WOODSON

Special Theme 2 - DISASTER INFORMATION

Natural-Technologic Events: The Frequency and Severity of Toxic Releases During and After

Natural Disasters with Emphasis on Wind and Seismic Events 261

Stacy YOUNG

Disaster Information Systems and Seismic Countermeasures for Field Facilities in NTT 267

Kazuhiko FUJIHASHI

Special Theme 3- MANAGEMENT OF INFRASTRUCTURE SYSTEMS

Management ofInfrastructure Facilities as Disaster Protection 283

Tadayuki TAZAKI, Takeo NAKAJIMA, and Kazuhiro NISHIKAWA

Information Systems for Infrastructure Management and Emergency Response 297

William E. ROPER

Risk Assessment on Storm Surge Flood 315

Kenichi TORII and Fuminori KATO

MANUSCRIPTS AUTHORED FOR PANEL MEETING BUT NOT PRESENTED

ORALLY

Shaking Table Tests on Seismic Response Reduction Effects ofRocking Building Structural

Systems 325

Tatsuya AZUHATA, Mitsumasa MIDORKAWA, Tadashi ISHIHARA, and Akira

WADA

Large-scale Tests for Performance Verification ofSmart Structures 333

Mitsumasa Midorikawa, Masaomi TESHIGAWARA, Masanori IIBA, Hideo FUJTTANI,

Koichi MORITA, Tatsuya AZUHATA, Akiyoshi FUKUDA, Yoichi SHIOZAKI, Takeshi

HIWATSHI, and Chikahiro MINOWA

Earthquake Recording in Buildings as a Toolfor the Comprehensive Structural

Performance Evaluation 339

Izuru OKAWA, Toshifumi FUKUTA, Masanori IIBA, Masaomi TESHIGAWARA,

Toshihide KASHIMA, Kouichi KUSUNOKI, and Shin KOYAM

APPENDIX

Task Committee Reports 349

XV

AGENDA

34th TECHNICAL PANEL MEETING on

WIND AND SEISMIC EFFECTS

13-15 May 2002

Monday 13 May

1000 OPENING CEREMONIES

(Lecture Room B, Administration Building)

Call to order by Mr. Stephen CAUFFMAN, Secretary-General US-side Panel

Opening remarks by Dr. Arden L. BEMENT, Jr. Director, National Institute of Standards

and Technology

Remarks by Mr. Koichi MORIMOTO, Counsellor for Science and Technology, Embassy

ofJapan

Remarks by Dr. Shyam SUNDER, Chairman US-side, Panel on Wind and Seismic Effects,

Chief, Structures Division, Building and Fire Research Laboratory

Remarks by Dr. Tadahiko SAKAMOTO, Chairman Japan-Side, Panel on Wind and

Seismic Effects, Chief Executive, Public Works Research Institute

Introduction of U.S. Members by U.S. Panel Chairman

Introduction ofJapan Members byJapan Panel Chairman

Elect Joint Meeting Chairman

Adopt Agenda

Adjourn

1115 Group Photograph

TECHNICAL SESSION 1

1 130-1250 Technical Session 1 - GEOTECHNICAL ENGINEERING AND

GROUND MOTION

Chairman: Dr. Tadahiko SAKAMOTO

1130 Seismic Performance of Urban, Reclaimed, and Port Areas - Full Scale

Experiment Using Blast Technique, Takahiro SUGANO* and Eiji

KOHAMA, PARI

1 1 45 A Review ofEmpirical Evidence For Site Coefficients in Building-Code Provisions,

xvii
Roger BORCHERDT, USGS (Presentation by Mehmet CELEBI)

1200 A Practical Procedure for Formulating Level 2 Earthquake Motions Based

on Scenario Earthquakes, Shojiro KATAOKA*, NILIM, Jun

MURAKOSHI, PWRI, and Keiichi TAMURA, PWRI

1215 Identifying Site Frequencies from Building Response Records, Mehmet CELEBI,

USGS

1230 Evaluation ofEarth Pressure Acting on Embedded Pile-cap Based on

Liquefaction Test Using Large-scale Shaking Table, Shuji TAMURA*,

Shinshu University, Chikahiro MINOWA, NRIESDP, Kohji TOKIMATSU,

TIT, Shunji FUJII, Kajima Tech. Res. Inst., Kaeko YAHATA, Taisei Corp.,

and Tomio TSUCHIYA, Takenaka Research & Dev. Inst. 1245 Discussion

1305 Lunch

TASK COMMITTEE MEETINGS

1415-1700 Task Committee Meetings

T/C A: Geotechnical Engineering and Ground Motion. Dr. Osamu MATUO, NILIM

and Dr. Takahiro SUGANO, PARI (Japan-side Co-Chairs) and Dr. Mehmet

CELEBI, USGS and Dr. Michael SHARP, Corps of Engineers (U.S.-side CoChairs)

Admin. Bldg. Room CR-111

T/C B: Next-generation Building and Infrastructure Systems. Dr. Masaomi

TESHIGAWARA, BRI and Mr. Keiichi OHTANI, NIED (Japan-side Co-Chairs)

and Dr. George LEE, MCEER and Dr. Peter CHANG, NSF (U.S.-side Co-Chairs)

Admin. Bldg. Room CR-113

T/C C: Dams. Dr. Yoshikazu YAMAGUCHI, PWRI (Japan-side Chair) and Dr. Robert

HALL, Corps of Engineers (U.S.-side Chair)

BFRL Conf. Room B-221 (Bldg 226)

T/C D: Wind. Dr. Hisashi OKADA, BRI and Dr. Nobuyuki HIRAHARA, PWRI (Japanside Co-Chairs) Dr. Peter CHANG, NSF (U.S.-side Chair)

Structures Conference Room B-139 (Bldg 226)

T/C G: Transportation Systems. Dr. Hiroshi SATO, PWRI Qapan-side Chair) and Dr.

Phillip YEN, FHWA (U.S.-side Chair)

Admin. Bldg. Lecture Room B

1700 Conclusion of Day 1

xviii
Tuesday 14 May

TECHNICAL SESSION 2

0815-0935 Technical Session 2 - NEXT-GENERATION BUILDING AND

INFRASTRUCTURE SYSTEMS

Chairman: Dr. Shyam SUNDER

0815 Project on 3-D Full Scale Earthquake Testing Facility, Keiichi OHTANI*, Tsuneo

KATAYAMA, Heki SHIBATA and Nobuyuki OGAWA, NIED

0830 Smart Damping Technologies for Dynamic Hazard Mitigation, B. F. Spencer Jr.

University of Notre Dame

0845 Research and Development Programs on Timber Structures in Japan, Hiroshi

ISODA, BRI, Hisashi OKADA, BRI, Naohito KAWAI, NILIM, and Nobuyoshi

YAMAGUCHI, PWRI Presented by Masaomi TESHIGAWARA, BRI

0900 Modeling Regional Economic Resilience to Earthquakes: A Computable general

Equilibrium Analysis ofLifeline Disruptions, Adam ROSE*, The Pennsylvania State

University and Shu-Yi LIAO, California Energy Commission

0915 Discussion

0935 Break

TECHNICAL SESSION 3

0955-1 1 15 Technical Session 3 - WIND ENGINEERING

Chairman: Dr. Shyam SUNDER

0955 Aerodynamic Stability ofa Super Long-Span Bridge with Slotted Box Girder,

Hiroshi SATO*, PWRI, Nobuyuki HIRAHARA, HSBA, Koichiro FUMOTO,

HSBA, Shigeru HIRANO, HSBA, and Shigeki KUSUHARA, HSBA

1010 Wind Effects on Long Span Cable Stayed Bridges: Assessment and Validation, Nicholas

JONES* and Ender OZKAN, Johns Hopkins University

1025 Strategy on Wind Engineering Research in BRI and NILIM, Hisashi OKADA, BRI,

Yasuo OKUDA* BRI, Hitomitsu KIKITSU, BRI and Masamiki OHASHI, NILIM

1040 Next Generation Wind Tunnels for Simulation ofStraight-Line, Thunderstorm- and

Tornado-Like Winds, Partha SARKAR*, and Fred L. HAAN, Jr., Iowa State University

1055 Discussion

1115 Adjourn For Lunch

1130 Lunch

TECHNICAL SESSION 4

1230-1335 Technical Session 4 - DAMS

Chairman: Dr. Shyam SUNDER

1230 Seismic Performance Evaluation ofIntake Towers, Richard C. DOVE*, US Army Corps

of Engineers and Enrique E. MATHEU, Louisiana State University

xix
1245 Present State ofMeasurement ofEarthquake Motion and Nation-Wide Networking

ofSeismographs at Dam Sites in Japan, Shuji TAKASU, PWRI, Hitoshi

YOSHIDA, NILIM, Yoshikazu YAMAGUCHP, PWRI, Takashi SASAKI, PWRI,

and Kenji INAGAKI, NILIM

1300 Deformation Analysis and Remediation for Small Earth Dams, Michael K. SHARP*, US

Army CORPS and Scott STEEDMAN, Whitby Byrd and Associates, London, UK

1315 Discussion

1335 Break

TECHNICAL SESSION 5

1355 - 1630 Technical Session 5 - TRANSPORTATION SYSTEMS

Chairman: Dr. Shyam SUNDER

1355 An Evaluation ofthe Force Reduction Factor in the Force-Based Seismic Design,

Gakuho WATANABE and Kazuhiko KAWASHIMA*, Tokyo Institute of

Technology

1410 Bolu Viaduct-1 Subjected to Near-Fault Ground Motion, Sunwoo PARK, Lendis

Corporation, Hamid GHASEMI*, FHWA, Jerry SHEN, Lendis Corporation, and Phillip

YEN, FHWA

1425 2002 Seismic Design Specifications for Highway Bridges, Shigeki UNJOH*, PWRI,

Shoichi NAKATANI, NILIM, Keiichi TAMURA, PWRI, Jiro FUKUI, PWRI, and

Jun-ichi HOSHIKUMA, PWRI

1 440 Development ofNational Guidelines for Seismic Performance Testing, W. Phillip YEN*

,

Jerry J. SHEN, and John O'FALLON, FHWA

1455 Discussion

STRATEGIC PLANNING SESSION

1530-1700 Strategic Planning Session (Panel Chairs and Secretariat)

1 700 Conclusion of Day 2

Wednesday 15 May

SPECIAL THEME 1 - TERRORISM: COLLAPSE of the WTC and

PENTAGON WEDGE A

Chairman: Dr. Tadahiko SAKAMOTO

0845-0950

0845 Standards, Technologies, and Practices for Cost-Effective Safety and Security ofBuildings

and Critical Facilities, Shyam SUNDER, NIST

0900 FEMA/ASCE Building Performance Assessment Team Study ofthe World Trade Center

Events, Michael MAHONEY, FEMA

XX
0915 Terrorist Threats Against the Pentagon, Robert L. HALL* , James T. BAYLOT, Robert J. DINAN, Richard DOVE, John HAYES, John HAYS, W. Matthew HOSSLEY, James L. O'DANIEL, Thomas R. SLAWSON, and Stanley C. WOODSON, US Army Corps of

Engineers

0930 Discussion

0950 Break

SPECIAL THEME 2 - DISASTER INFORMATION

Chairman: Dr. Tadahiko SAKAMOTO

1005-1050

1 005 Natural- Technologic Events: The Frequency And Severity ofToxic Releases During and

After Natural Disasters With Emphasis on Wind and Seismic Events, Stacy YOUNG,

National Center for Environmental Health, CDC

1020 Disaster Information Systems and Seismic Countermeasure for Field Facilities in

NTT, Kazuhiko FUJIHASHI, NTT

1035 Discussion

1050 Break

SPECIAL THEME 3- MANAGEMENT OF INFRASTRUCTURE SYSTEMS

Chairman: Dr. Tadahiko SAKAMOTO

1210

Management ofInfrastructure Facilities to Ensure the Function as Disaster

Protection, Tadayuki TAZAKI, Takeo NAKAJIMA, and Kazuhiro NISHIKAWA*,

NILIM

Information Systems for Infrastructure Management and Emergency Response, William

ROPER, George Washington University

Risk Assessment on Storm Surge Flood, Ken-ichi TORII* and Fuminori KATO,

NILIM

Discussion

Adjourn for lunch

Lunch

TASK COMMITTEE REPORTS AND RESOLUTIONS

1 330-1 430 Report of Task Committees

Chairman: Dr. Tadahiko SAKAMOTO

T/C A: Geotechnical Engineering and Ground Motion

T/C B: Next-generauon Building and Infrastructure Systems

T/C C: Dams

1105-

1105

1120

1135

1150

1210

1215

xxi
T/C D. Wind Engineering

T/C G: Transportation Systems and summary of 17"1 US-Japan Bridge Workshop, November 2001

1430-1530 Adoption of Final Resolutions

1530 Break

CLOSING CEREMONIES

1545 Call to Order by Mr. Stephen CAUFFMAN, Secretary-General, U.S.-Side Panel

Closing Remarks by Dr. Tadahiko SAKAMOTO, Chairman Japan-Side Panel

Closing Remarks by Dr. Shyam SUNDER, Chairman U.S.-Side Panel

1615 Conclusion of 34th Joint Panel Technical Sessions

xxii
LIST OF PANEL MEMBERS

UNITED STATES-SIDE PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

30 April 2002

Dr. S. Shyam Sunder

Chairman, U.S.-side Panel

Chief, Structures Division

Building and Fire Research Laboratory

National Institute of Standards and

Technology

100 Bureau Drive, Stop 861

Gaithersburg, MD 20899-861

301-975-6061 FAX:301 -869-6275

sunder@nist.gov

Mr. Stephen A. Cauffman

Secretary-General, U.S.-side Panel

Research Engineer

National Institute of Standards and

Technology

Building and Fire Research Laboratory,

Structures Division

100 Bureau Drive, Stop 861

Gaithersburg, MD 20899-8611

301-975-6051 FAX: 301-869-6275

stephen.cauffman@nist.gov

Dr. Daniel P. Abrams

Director, Mid-America Earthquake Center

Hanson Engineers Professor of Civil

Engineering

University of Illinois

1241 Newmark Laboratory

205 North Mathews Avenue

Urbana, IL 61801

217-244-6302 FAX:217-333-3821

d-abrams@uiuc.edu

Dr. Kharaiti L. Abrol

Principal Structural Engineering Consultant

Room 475

Department of Veterans Affairs

810 Vermont Avenue, NW

Washington, DC 20420

202-565-5579 FAX:202-565-5478

abrkha@hq .med. va. gov

Dr. Jon Ake

Geophysicist

Seismotectonics and Geophysics Section

CodeD-8330

P.O. Box 25007

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303-445-3169 FAX:303-236-9127

i ake@do-usbr. gov

Dr. Clifford J. Astill

Program Director

Division of Biological and Critical Systems

National Science Foundation

4201 Wilson Boulevard - Room 545

Arlington, VA 22230

703-306-1361 FAX:703-306-0291

castill@nsf.gov

Mr. John Baals

Seismic Safety Coordinator

P.O. Box 25007 (D-8110)

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303-236-3999 x534 FAX:303-236-9099

ibaals@do.usbr.gov

Dr. Celso S. Barrientos

Supervisory Physical Scientist

National Environmental Satellite Data

Information Service - Code E/RA28

National Oceanic and Atmospheric

Administration

U.S. Department of Commerce

5200 Auth Road

XXV
Camp Springs, MD 20746

301-713-9384x109 FAX: 301-713-4598

Celso.S.Barrientos@noaa.gov

Dr. Eddie N. Bernard

Director, Pacific Marine Environmental

Laboratory

National Oceanic and Atmospheric

Administration

U.S. Department of Commerce

7600 Sand Point Way, NE

BINC15700/Building3

Seattle, WA 98115-0070

206-526-6800 FAX:206-526-4576

Eddie.N.Bemard@noaa.gov

Dr. David M. Boore

U.S. Geological Survey

345 Middlefield Road MS 977

Menlo Park, CA 94025

650-329-5616 FAX:650-329-5163

boore@usgs.gov

Dr. Roger D. Borcherdt

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

650-329-5619 FAX:650-329-5163

borcherdt@usgs. gov

Dr. Mehmet K. Celebi

Research Civil Engineer

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

650-329-5623 FAX: 650-329-5163

celebi@usgs.gov

Mr. Harish Chander

U.S. Dept. of Energy (EH-31),

CXXLRoom 2016(GTN)

19901 Germantown Road

Germantown, MD 20874-1290

301-903-6681 FAX:301 -903-8693

harish. chander@hq. doe. gov

Dr. Peter Chang

National Science Foundation

4201 Wilson Boulevard - Room 545

Arlington, VA 22230

703-292-8360

pchang@nsf.gov

Dr. C.Y. Chen

Senior Civil/Geotechnical Engineer

Office of Foreign Buildings

Department of State

Code SA-6, Rm. 347

Washington, DC 20520

703-875-6207 FAX:703-875-6204

chency@state.gov

Dr. Ken Chong

(Serving as Guest Researcher NIST's

Building Materials Division during FY

2002)

Program Director

Mechanics and Materials

Engineering Directorate

National Science Foundation

4201 Wilson Boulevard - Room 545

Arlington, VA 22230

703-292-7008 FAX:703-292-9053

kchong@nsf.gov

Mr. James D. Cooper

Director, Office of Bridge Technology

Federal Highway Administration

U.S. Department of Transportation

400 7

th Street SW

Washington, DC

202-366-4589 FAX 703-285-2766

Jim.Cooper@igate.fhwa.dot.gov

Dr. James F. Costello

Senior Structural Engineer

Structural and Geological Engineering

Branch

xxvi
Office of Nuclear Regulatory Research

US Nuclear Regulatory Commission

Mail Stop T10-L1

Washington, DC 20555

301-415-6009

ifc@nrc.gov

Mr. William Freeborne

Housing and Urban Development

Room 8132

451 Seventh Street, SW

Washington, DC 20410

202-708-4370 x5725 FAX:202-708-5873

william_e._fi'eebome@hud.gov

Dr. Joseph Golden

Forecast Systems Laboratory

National Oceanic and Atmospheric

Administration

Rm. DSRC-2B133

325 Broadway Street

Boulder, CO 80302

303-497-6098

j golden@fsl.noaa. gov

Dr. John Gross

Leader, Structural Systems and Design

Group

Building and Fire Research Laboratory

National Institute of Standards and

Technology

100 Bureau Drive, Stop 861

Gaithersburg, MD 20899-861

301-975-6068 FAX:301 -869-6275

iolm.gross@nist.gov

Dr. Robert L. Hall

Chief, Structural Analysis Group

US Army Engineer Research and

Development Center

Waterways Experiment Station

Office Address: CEERD-GS

3909 Halls Ferry Road

Vicksburg,MS 39180-6199

601-634-2567 FAX:601-634-3412

Robert.L.Hall@erdc. usace.army.mil

Dr. Allen M. Hittelman

Chief, Solid Earth Geophysics Division

E/GC1, National Geophysical Data Center

NOAA, NESDIS

325 Broadway

Boulder, CO 80303-3328

303-497-6591 FAX:303-497-6513

all en .m.hittelman@noaa. gov

Dr. Mary Ellen Hynes

Chief, Earthquake Engineering and

Geophysics Branch

Geotechnical Laboratory

Office Address: CEERD-GS-R

US Army Engineer Research and

Development Center

Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180

601-634-2280 FAX:601-634-3453

mary.e.hynes@erdc.usace.army.mil

Mr. Roger M. Kenneally

Structural Engineer

Structural and Seismic Engineering Branch

Mail Stop T10-L1

US Nuclear Regulatory Commission

Washington, DC 20555

301-415-6303

rmk@nrc.gov

Mr. James Lander

Geophysicist

Cooperative Institute for Research in

Environmental Sciences

University of Colorado

Campus Box 449, Room 152 RL3

3100 Marine Street

xxvii
Boulder, CO 80309

303-497-6446 FAX: 303-497-6513

ilander@ngdc.noaa.gov

Dr. George Lee

Director, Multidisciplinary Center for

Earthquake Engineering Research

(MCEER)

State University of New York at Buffalo

Red Jacket Quadrangle

Buffalo, NY 14261

716-645-3391 xlll FAX:716-645-3399

gclee@mceermail.buffalo.edu

Dr. H.S. Lew

Structures Division

Building and Fire Research Laboratory

National Institute of Standards and

Technology

100 Bureau Drive Stop 861

1

Gaithersburg, MD 20899-861

301-975-6060 FAX:301 -869-6275

hai.lew@nist.gov

Mr. Michael Mahoney

Physical Scientist

Federal Emergency Management Agency

500 C Street, SW

Washington, DC 20472

202-646-2794 FAX:202-646-4387

mike.mahonev@fema.gov

Dr. Josephine Malilay

Disaster Assessment and Epidemiology

Section

National Center for Environmental Health

F-46 Center for Disease Control and

Prevention

4770 Buford Highway, NE

Atlanta, GA 30341

404-489-1359 FAX: 404-498-1355

imalilay@cdc.gov

Dr. Francis G. McLean

Value Engineering Program Manager

Code D-8 170

P.O. Box 25007

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303-445-3091 FAX:303-6475

FMCLEAN@do.usbr. gov

Dr. Raymond E. Meyer

Department of State

SA14/Room 1705

USAID/Washington, DC 20523-1443

202-712-1078 FAX: 202-216-3707

Rmeyer@usaid.gov

Dr. Martin C. Miller

Chief, Coastal Oceanography Branch

USAE, Waterways Experiment Station

Coastal and Hydraulics Laboratory

(CEWES-CR-O)

3909 Halls Ferry Road

Vicksburg, MS 39180

601-634-3999 FAX:601-634-4314

martin.c.miller@erdc.usace.aimy.mil

Dr. Jack Moehle

Director, Pacific Earthquake Engineering

Research Center (PEER)

1301 South 46th Street

Richmond, CA 94804-4698

510-231-9554 FAX: 510-231-9471

moehle@eerc.berkeley.edu

Mr. Ugo Morelli

Policy Manager

Federal Emergency Management Agency

500 C Street, SW

Washington, DC 20472

202-646-2810 FAX:202-646-2577

ugo.morelli@fema.gov

Dr. William Roper

Professor and Chairman

xxviii
Civil and Environmental

Engineering Department

George Washington University

Phillips Hall, Room 643

The Academic Center

801 22nd St., N.W.

Washington, DC 20052

202-994-4901 FAX: 202 994-0127

wroper@seas. gwu.edu

Dr. Erdal Safak

Research Structural Engineer

U.S. Geological Survey

Box 25046, MS 966

Denver Federal Center

Denver, CO 80225

303-273-8593 FAX:303-273-8600

safak@usgs.gov

Dr. Michael Sharp

Director, Centrifuge Research Center

Geotechnical and Structures Laboratory

US Army Engineer Research and

Development Center

Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-4127 FAX: 601-634-4894

michael.k.sharp@erdc.usace.ai7ny.mil

Dr. Charles E. Smith

Senior Technical Advisor

Engineering and Research Branch

U. S. Minerals Management Service

381 Elden Street, MS 4021

Herndon, VA 20170-4817

703-787-1561 Fax 703-787-1549

smithc@mms.gov or Charles.E.Smith@mms.gov

Dr. W. Phillip Yen

Seismic Hazard Mitigation

Office of Infrastructure, R&D

Federal Highway Administration

6300 Georgetown Pike

McLean, VA 22101

202-493-3056 FAX: 202-493-3442

Wen-huei.Yen@fhwa.dot.gov

Mr. Ray Zelinski

Chief, Earthquake Engineering Office

California Department of Transportation

Division of Engineering Services, MS9-2/7J

P.O.Box 168041

Sacramento, CA 95816

916-227-8806 FAX: 916-227-8242

ray.j.zelinski@dot.ca.gov

xxix
TEMPORARY US-SIDE MEMBERS

Dr. Richard C. Dove

US Army Engineer Research and

Development Center

Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg. MS 39180-6199

601-634-2883

richard.c.dove@erdc.usace.army.mil

Dr. Hamid Ghasemi

Turner-Fairbank Highway Research Center

Federal Highway Administration

6300 Georgetown Pike

McLean, VA 22101

703-285-2447

hamid.ghasemi@fhwa.dot.gov

Dr. Nicholas P. Jones

Professor and Chair

Department of Civil Engineering

Johns Hopkins University

3400 N. Charles Street

Baltimore, MD 21218-2686

410-516-7874; (F) 410-516-7473

nick@jhu.edu

Dr. Enrique E. Matheu

Department of Civil and Environmental

Engineering

Louisiana State University

Office address:

U.S. Army Engineer Research and

Development Center

3909 Halls Ferry Road, ATTN: CEERD-GSM

Vicksburg, MS 39180

Tel: 601-634-3235, Fax: 601-634-3412

Enrique.E.Matheu@erdc.usace.army.mil

Professor Adam Rose

Department Head, Energy, Environmental, &

Mineral Economics

The Pennsylvania State University

221 Eric A. Walker Building

University Park, PA 16802-5010

(T) 814-863-0179

azrl@psu.edu

Dr. Partha P. Sarkar

Associate Professor and Wilson Chair

Director, Wind Simulation and Testing

Laboratory

Department of Aerospace Engineering and

Engineering Mechanics

2271 Howe Hall, Room 1200

Iowa State University

Ames, LA 5001 1-2271

515-294-0719 515-294-3262

ppsarkar@iastate.edu

Dr. B.F. Spencer, Jr. Leo E. and Patti Ruth Linbeck Professor of

Engineering

Department of Civil Engineering and

Geological Sciences

156 Fitzpatrick Hall

University of Notre Dame

Notre Dame, IN 46556-0767

(T) 219-631-6247, (F) 443-646-0675

219-315-0738 (Mobile)

spencer@nd.edu

Stacy Young, MS MPH

Fellow, Association of Schools of Public

Health

Centers for Disease Control and Prevention

National Center for Environmental Health

1600 Clifton Rd., NE (Mailstop E-23)

Atlanta, GA 30333

404-498-1877; 404-498-1355

say5@cdc.gov

XXX
JAPAN-SIDE PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

STEERING COMMITTEE MEMBERS

Dr. Tadahiko Sakamoto

Chairman, Japan-side Panel on Wind and Seismic Effects

Chief Executive, Independent Administrative Institution

Public Works Research Institute

1-6 Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6700

Fax: 0298-79-6741

E-mail: sakamoto@pwri.go.jp

Dr. Michio Okahara

Secretary-General, Japan-side Panel on Wind and Seismic Effects

Executive Director, Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6700

Fax: 0298-79-6702

E-mail: okahara@pwri.go.jp

Mr. Minoru Akiyama

Director, Geographic Department

Geographical Survey Institute

Ministry of Land, Infrastructure and Transport

1, Kitasato, Tsukuba-shi, Ibaraki-ken 305-081

1

Tel:0298-64-llll

Fax:0298-64-1804

E-mail: m-akiyoama@gsi.go.jp

Mr. Ippei Eguchi

Senior Assistant for Disaster Prevention

Planning Division

Administration Department

Japan Meteorological Agency

1-3-4, Ohte-machi, Chiyoda-ku, Tokyo 100-8918

Tel: 03-3212-8341

Fax: 03-3211-2032

Mr. Tadayuki Gotou

Director, Building Disaster Prevention Office

Housing Bureau

Ministry of Land, Infrastructure and Transport

2- 1-3, Kasumigaseki, Chiyoda-ku, Tokyo 100-8918

Tel: 03-5253-8514

Fax: 03-5253-1630

E-mail: gotou-t2gr@mlit.go.jp

xxxi
Dr. Nobuo Hamada

Director, Seismological Volcano logicale

Research Department

Meteorological Research Institute

Japan Meteorological Agency

1- 1, Nagamine, Tsukuba-shi, Ibaraki-ken 305-0052

Tel: 0298-53-8675

Fax: 0298-53-3730

E-mail: nhamada@mri-jma.go.jp

Dr. Susumu Iai

Director, Earthquake Disaster Prevention Laboratory

Independent Administrative Institution

Port and Airport Research Institute

3- 1 -

1 , Nagase, Yokosuka-shi, Kanagawa-ken 239-0826

Tel: 0468-44-5030

Fax: 0468-44-0839

E-mail: iai@ipc.pari.go.jp

Mr. Masanori Iiba

Chief Research Engineer

Department of Structural Engineering

Independent Administrative Institution

Building Research Institute

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-8020

Tel:0298-64-6637

Fax:0298-64-6770

E-mail: iiba@kenken.go.jp

Mr. Takashi Kaminosono

Research Coordinator for Disaster Mitigation of Buildings

Research Center for Disaster Risk Management

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel:0298-64-4963

Fax:0298-64-0598

E-mail: kaminosono-t92rx@nilim.go.jp

Prof. Kazuhiko Kawashima

Professor, Department of Civil Engineering

Tokyo Institute of Technology

2- 12-1, O-okayama, Meguro-ku, Tokyo 152-0033

Tel: 03-5734-2922

Fax: 03-3729-0728

E-mail: kawasima@cv.titech.ac.jp

xxxii
Mr. Seiya Kinoshita

Director, International Division for Infrastructure

Policy Bureau

Ministry of Land Infrastructure and Transport

2-1-3, Kasumigaseki, Chiyoda-ku, Tokyo 100-8918

Tel: 03-5253-8315

Fax:03-5253-1562

E-mail: kinoshita-s2jg@mlit.go.jp

Mr. Satoru Kondo

Director, River Division

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel:0298-64-2211

Fax:0298-64-1168

E-mail: kondou-s86uj@nilim.go.jp

Mr. Norishige Kubo

Deputy Director, International Division for Infrastructure

Policy Bureau

Ministry of Land Infrastructure and Transport

2-1-3, Kasumigaseki, Chiyoda-ku, Tokyo 100-8918

Tel:03-5253-8315

Fax:03-5253-1562

Mr. Takaaki Kusakabe

Head, Earthquake Disaster Prevention Division

Research Center for Disaster Risk Management

Institute for Land and Infrastructure and Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

0298-64-2211

0298-64-0598

E-mail: ksakabe-t88d8@nilim.go.jp

Mr. Osamu Matsuo

Research Coordinator for Earthquake Disaster Prevention

Research Center for Disaster Risk Management

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2926

Fax: 0298-64-0598

E-mail: matsuo-o92rm@nilim.go.jp

xxxiii
16 Mr. Kiyoshi Minami

Director, Division of Road Management

Road Bureau

Ministry of Land, Infrastructure and Transport

2- 1-3, Kasumigaseki, Chiyoda-ku, Tokyo 100-8918

Tel: 03-5253-8111

Fax: 03-5253-1618

E-mail: minami-k2z4@mlit.go.jp

Mr. Chikahiro Minowa

Scientific Research Advisor

Independent Administrative Institution

Research Institute for Earth Science and Disaster Prevention

3- 1, Tennodai, Tsukuba-shi, Ibaraki-ken 305-0006

Tel: 0298-51-1611

Fax: 0298-51-1674

E-mail: minowa@bosai.go.jp

Mr. Masayuki Mori

International Cooperation Coordinator,

International Science and Technology Affairs Division,

Science and Technology Policy Bureau

Ministry of Education, Culture, Sports, Science and Technology

2-2-1, Kasumigaseki, Chiyoda-ku, Tokyo 100-8918

Tel: 03-5253-4055

Fax: 03-3581-5909

E-mail: m5mori@mext.go.jp

Mr. Akiyoshi Mukai

Chief Research Engineer, Department of Structural Engineering

Independent Administrative Institution

Building Research Institute

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel:0298-64-6662

Fax:0298-64-6773

E-mail: mukai@kenken.go.jp

Mr. Jun Murakoshi

Team Leader, Bridge Structure Team,

Structures Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel:0298-79-6793

Tel:0298-79-6739

E-mail: murakosi@pwri.go.jp

xxxiv
Dr. Isao Nishiyama

Advanced Research Engineer

Production Engineering Department

Independent Administrative Institution

Building Research Institute

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6681

Fax: 0298-64-6772

E-mail: isao@kenken.go.jp

Mr. Akihiko Nunomura

Counselor .Prime Ministers' Secretarist

Cabinet Office

1-2-2, Kasumigaseki, Chiyoda-ku, Tokyo 100-8918

Tel: 03-3501-5693

Fax: 03-3501-5199

Dr. Hisashi Okada

Director, Structural Engineering Department

Independent Administrative Institution

Building Research Institute

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-2151

Fax: 0298-64-2989

E-mail: okada@kenken.go.jp

Mr. Kazunori Oodaira

Director, Disaster Risk Management Division

River Bureau

Ministry of Land, Infrastructure and Transport

2- 1-3, Kasumigaseki, Chiyoda-ku, Tokyo 100-8918

Tel: 03-3580-8111

Fax: 03-5253-1608

E-mail: oodaira-k2rb@mlit.go.jp

Dr. Izuru Ohkawa

Head, Construction Techniques Division

Production Department

Building Research Institute

Independent Administrative Institution

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6627

Fax: 0298-64-6773

E-mail: okawa@kenken.go.jp

XXXV
Mr. Keiichi Ohtani

Project-Director, Project of 3-D Full-Scale Earthquake Testing Facility

National Research Institute for Earth Science and Disaster Prevention

3-1, Tennodai, Tsukuba-shi, Ibaraki-ken 305-0006

Tel: 0298-51-1611

Fax: 0298-52-8512

E-mail: ohtani@bosai.go.jp

Mr. Takuaki Ohyama

Head, Infections Diseases Division

Institute of Infections Diseases

Ministry of Health, Labor and Welfare

1-23-1, Toyama, Shinjyuku-ku, Tokyo 162-8640

Tel: 03-5285-1268

Fax: 03-5285-1129

E-mail: Ohyamat@nih.go.jp

Dr. Hiroshi Sato

Director, Structures Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6726

Fax: 0298-79-6739

E-mail: hsato@pwri.go.jp

Dr. Takahiro Sugano

Chief, Structural Dynamics Laboratory

Structural Engineering Division

Independent Administrative Institution

Port and Airport Research Institute

3-1-1, Nagase, Yokosuka-shi, Kanagawa-ken 239-0826

Tel: 0468-44-5058

Fax: 0468-44-0839

E-mail: macsuga@ipc.pari.go.jp

Mr. Tetsuya Suzuki

Director, Structure Department

Independent Administrative Institution

Civil Engineering Research Institute of Hokkaido

1-3, Hiragishi, Toyohira-ku, Sapporo-shi

Hokkaido 062-8601

Tel: 011-841-5175

Fax: 011-820-2714

E-mail: 73302@ceri.go.jp

xxxvi
Mr. Yasushi Taguchi

Director, Research and Development Policy Division

Research and Development Bureau

Ministry of Education, Culture, Sports ,Science and Technology

l-3-2,Kasumigaseki,Chiyoda-ku,Tokyo 100-8966

Tel: 03-5253-4134

Fax: 03-5253-4139

E-mail: ytaguchi@mext.go.jp

Dr. Shigeo Takahashi

Director, Hydraulic Engineering Division

Independent Administrative Institution

Port and Harbor Research Institute

3-1-1, Nagase, Yokosuka-shi, Kanagawa-ken 239-0826

Tel: 0468-44-5036

Fax: 0468-44-1274

E-mail: takahashi_s@cc.pari.go.jp

Mr. Shuji Takasu

Director, Hydraulic Engineering Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6723

Fax: 0298-79-6737

E-mail: takasu@pwri.go.jp

Mr. Yukio Takemura

Director, Typhoon Research Division

Meteorological Research Institute

Japan Meteorological Agency

Ministry of Land, Infrastructure and Transport

1-1, Nagamine, Tsukuba-shi, Ibaraki-ken 305-0052

Tel: 0298-53-8663

Fax: 0298-53-8735

E-mail: ytakemur@mri-jma.go.jp

Dr. Keiichi Tamura

Team Leader, Ground Vibration Research Team

Earthquake Disaster Prevention Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6770

Fax: 0298-79-6735

E-mail: tamura@pwri.go.jp

xxxvii
Mr. Tadayuki Tazaki

Director-General, Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel:0298-64-2211

Fax:0298-64-2148

E-mail tazaki-t82aa@nilim.go.jp

Dr. Masaomi Teshigawara

Chief Research Engineer

Department of Structural Engineering, Research Coodinator

Independent Administrative Institution

Building Research Institute

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6753

Fax: 0298-64-6773

E-mail: teshi@kenken.go.jp

Mr. Ken-ichi Tokida

Director, Earthquake Disaster Prevention Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6700

Fax:0298-79-6736

E-mail: tokida@pwri.go.jp

Mr. Ken-ichi Torii

Head, Coastal Division

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2327

Fax: 0298-64-1168

E-mail: torii-k92fr@nilim.go.jp

Mr. Shinichiro Ueno

Director, Disaster Prevention Office

Coast Administration and Disaster Prevention Division

Ports and Harbors Bureau

Ministry of Land, Infrastructure and Transport

2- 1-3, Kasumigaseki, Chiyoda-ku, Tokyo 100-8918

Tel: 03-5253-8689

Fax: 03-5253-1654

E-mail: ueno-s2sk@milt.go.jp

xxxviii
Dr. Shigeki Unjoh

Team Leader, Earthquake Engineering Research Team

Earthquake Disaster Prevention Research Group

Independent Administrative Institution

Public Works Research Institute

Minamiharal-6, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6773

Fax: 0298-79-6736

E-mail: unjoh@pwri.go.jp

Dr. Yoshikazu Yamaguchi

Team Leader, Dam Structure Research Team

Hydraulic Engineering Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6780

Fax: 0298-79-6737

E-mail: yamaguti@pwri.go.jp

Dr. Hiroyuki Yamanouchi

Chief Executive

Independent Administrative Institution

Building Research Institute

1, Takehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-2151

Fax: 0298-64-2989

E-mail: yamanoch@kenken.go.jp

xxxix
TEMPORARY COMMITTEE MEMBERS

Mr. Kazuhiko Fujihashi

Senior Research Engineer, Supervisor

Disaster Prevention and Environmental Protection Group,

Civil Engineering Project

Access Network Service Systems Laboratories

Nippon Telegraph and Telephone Corporation (NTT)

1-7-1, Hanabatake, Tsukuba-shi, Ibaraki-ken 305-0805

Tel: 0298-68-6241

Fax: 0298-68-6259

E-mail: fuj iha si @ansl .ntt. co.jp

Mr. Koichiro Fumoto

Senior Researcher, Structures Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6726

Fax: 0298-79-6739

E-mail: fumoto@pwri . go. ip

Mr. Masahiro Ishida

Senior Researcher, Research Administration and International Cooperation Division

Planning and Research Administration Department

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2211

E-mail: ishida-m92bs@nilim.go.jp

Mr. Shojiro Kataoka

Senior Researcher

Earthquake Disaster Prevention Division

National Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-3245

Fax: 0298-64-0598

E-mail: kataoka-s92rc@nilim.go.jp

Mr. Takeo Nakajima

Director, Planning and Research Administration Department

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2826

Fax: 0298-64-1527

E-mail: nakajima-t92b4@nilim.go.jp

xl
Mr. Kazuhiko Nishikawa

Research Coordinator for evaluation

Planning and Research Administration Department

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-3725

Fax: 0298-64-1527

E-mail: nishikawa-k92br@nilim.go.jp

Dr. Shuji Tamura

Associate Professor, Architecture and Civil Engineering

Faculty of Engineering

Shinshu University

4-17-1, Wakasato, Nagano-shi 380-8553

Tel: 026-269-5362

E-mail: tamura@gipwc.shinshu-u.ac.jp

xli
SECRETARIES COMMITTEE MEMBERS

Dr. Michio Okahara

Secretary-General, Japan-side Panel on Wind and Seismic Effects

Executive Director, Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6700

Fax: 0298-79-6702

E-mail: okahara@pwri.go.jp

Mr. Jiro Fukui

TeamLeader, Foundation Engineering Team

Structure Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6794

Fax: 0298-79-6739

E-mail: fukui@pwri.go.jp

Mr. Kenji Ikeda

Director, Structures Division

Structures Department

Independent Administrative Institution

Civil Engineering Research Institute of Hokkaido

1-3, Hiragishi, Toyohira-ku, Sapporo-shi

Hokkaido 062-8602

Tel: 011-841-1115

Fax: 011-820-2714

E-mail: ikedak@ceri.go.jp

Mr. Naohito Kawai

Head, Structural Standard Division

Building Department

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Tachihara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-4299

Fax: 0298-64-6774

E-mail:kawai-n92hp@nilim.go.jp

Mr. Takaaki Kusakabe

Head, Earthquake Disaster Prevention Division

Research Center for Disaster Risk Management

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2211

Fax: 0298-64-0598

E-mail:ksakabe-t88d8@nilim.go.jp

xlii
Mr. Osamu Matsuo

Research Coordinator for Earthquake Disaster Prevention

Research Center for Disaster Risk Management

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2926

Fax: 0298-64-0598

E-mail: matsuo-o92rm@nilim.go.jp

Mr. Chikahiro Minowa

Scientific Research Advisor

Independent Administrative Institution

Research Institute for Earth Science and Disaster Prevention

3-1, Tennodai, Tsukuba-shi, Ibaraki-ken 305-0006

Tel: 0298-51-1611

Fax: 0298-51-1674

E-mail: minowa@bosai.go.jp

Mr. Jun Murakoshi

TeamLeader, Bridge Structure Team

Structures Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken, 305-8516

Tel: 0298-79-6793

Fax: 0298-79-6739

E-mail: murakosi@pwri.go.jp

Mr. Shoichi Nakatani

Head, Bridge Division

Road Department

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2905

Fax: 0298-64-0178

E-mail: nakatani-s92gw@nilim.go.jp

Dr. Tetsuo Nakazawa

Head, The Second Research Laboratory

Typhoon Research Department

Meteorological Research Institute

Japan Meteorological Agency

Ministry of Land, Infrastructure and Transport

1-1, Nagamine, Tsukuba-shi, Ibaraki-ken 305-0052

Tel: 0298-53-8671

Fax: 0298-53-8735

E-mail : nakazawa@mri-jra.go.jp

xliii
Dr. Nobuyuki Ogawa

Deputy Project Director,

Special Project on the 3-D Full-Scale

Earthquake Testing Facility

Disaster Prevention Research Group

National Research Institute for Earth Science and Disaster Prevention

3-1, Tennodai, Tsukuba-shi, Ibaraki-ken 305-0006

Tel: 0298-51-1611

Fax: 0298-52-8512

E-mail: ogawa@bosai.go.jp

Dr. Hisashi Okada

Director, Structural Engineering Department

Independent Administrative Institution

Building Research Institute

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-2151

Fax: 0298-64-2989

E-mail: okada@kenken.go.jp

Dr. Yasuo Okuda

Chief Research Engineer, Department of Structural Engineering

Independent Administrative Institution

Building Research Institute

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6618

Fax: 0298-64-6773

E-mail: y_okuda@kenken.go.jp

Mr. Keiichi Ohtani

Project-Director, Project of 3-D Full-Scale Earthquake Testing Facility

Independent Administrative Institution

National Research Institute for Earth Science and Disaster Prevention

3-1, Tennodai, Tsukuba-shi, Ibaraki-ken, 305-0006

Tel: 0298-51-1611 Ext. 321

Fax: 0298-52-8512

E-mail: ohtani@bosai.go.jp

Mr. Yoshiyuki Saito

General Projector for Research

Independent Administrative Institution

Civil Engineering Research Institute of Hokkaido

1-3, Hiragishi, Toyohira-ku, Sapporo-shi

Hokkaido 062-0931

Tel: 011-841-1119

Fax: 011-824-1226

E-mail: 90117@ceri.go.jp

xliv
Dr. Hiroshi Sato

Director, Structures Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6726

Fax: 0298-79-6739

E-mail: hsato@pwri.go.jp

Dr. Takahiro Sugano

Head, Structural Dynamics Division

Geotechnical and Structural Engineering Department

Independent Administrative Institution

Port and Airport Research Institute

3-1-1, Nagase, Yokosuka-shi, Kanagawa-ken 239-0826

Tel: 0468-44-5058

Fax: 0468-44-0839

E-mail: macsuga@ipc.pari.go.jp

Dr. Keiichi Tamura

TeamLeader, Ground Vibration Research Team

Earthquake Disaster Prevention Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6770

Fax: 0298-79-6735

E-mail: tamura@pwri.go.jp

Dr. Masaomi Teshigawara

Chief Research Engineer

Structural Engineering Department

Independent Administrative Institution

Building Research Institute

1, Tatehara, Tsukuba-shi, Ibaraki-ken 305-0802

Tel: 0298-64-6753

Fax: 0298-64-6773

E-mail: teshi@kenken.go.jp

Mr. Ken-ichi Tokida

Director, Earthquake Disaster Prevention Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6700 (ext.4211)

Fax: 0298-79-6736

E-mail: tokida@pwri.go.jp

xlv
Mr. Ken-ichi Torii

Head, Coastal Division

River Department

Institute for Land and Infrastructure Management

Ministry of Land, Infrastructure and Transport

1, Asahi, Tsukuba-shi, Ibaraki-ken 305-0804

Tel: 0298-64-2327

Fax: 0298-64-1168

E-mail: torii-k92fr@nilim.go.jp

Mr. Masaharu Tsuzawa

Head, Planning Division

Geographic Department

Geographical Survey Institute

Ministry of Land, Infrastructure and Transport

1, Kitasato, Tsukuba-shi, Ibaraki-ken 305-0811

Tel: 0298-64-5917

Fax: 0298-64-1804

E-mail: tsuzawa@gsi.go.jp

Dr. Shigeki Unjoh

TeamLeader, Earthquake Engineering Research Team

Earthquake Disaster Prevention Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6773

Fax: 0298-79-6736

E-mail: unjoh@pwri.go.jp

Dr. Yoshikazu Yamaguchi

TeamLeader, Dam Structure Research Team

Hydraulic Engineering Research Group

Independent Administrative Institution

Public Works Research Institute

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516

Tel: 0298-79-6780

Fax: 0298-79-6737

E-mail: yamaguti@pwri.go.jp

xlvi
RESOLUTIONS

May 15, 2002

RESOLUTIONS OF THE THIRTY-FOURTH JOINT MEETING

U.S.-JAPAN PANEL ON WIND AND SEISMIC EFFECTS (UJNR)

National Institute of Standards and Technology, Gaithersburg, Maryland, USA

13-15 May 2002

The following resolutions are hereby adopted:

1 . The Thirty-fourth Joint Panel Meeting provided the forum to exchange valuable technical

information that is beneficial to both countries. In view of the importance of cooperative

programs on the subject of wind and seismic effects, the continuation of Joint Panel

Meetings is considered essential. Both sides agreed to explore new ideas and areas to

strengthen cooperative activities in support of the Panel's mission.

2. The following activities have been conducted since the Thirty-third Joint Meeting:

a. Technology Exchanges. Technical experts, technical documents, and applications

of the electronic media have been exchanged. These exchanges have contributed

to the development of new research programs and enhanced ongoing research in

both countries.

b. Task Committee Workshops and meetings. The Panel held one workshop and one

meeting:

1. Task Committee (B), U.S.-Japan grantee's meeting for Composite and

Hybrid Structures, June 24-27, 2001, Berkeley CA, USA.

2. Task Committee (G), 17th U.S.- Japan Bridge Engineering Workshop, 12-

14 November 2001, Tsukuba, Japan.

3. The Panel approved the Task Committee reports presented during the 34th Joint Panel

Meeting. Each report included objectives, scope of work, accomplishments and future

plans.

4. The Panel's Strategic Plan was approved at its 33rd Joint Panel Meeting. Implementation

of the Strategic Plan has started. The following are the highlights of the 34th Joint

Meeting.

xlix
a. The full Panel Meetings were conducted within the new assembly format: threeday technical meetings followed by three-days of technical site visits rather than

the former 12-day full Panel Meetings.

b. The Strategic Plan encourages greater private sector involvement as was

demonstrated from papers and presentations from Universities and

Government/private sector partner experts. In addition, two of the three technical

site visits were hosted by private sector organizations.

c. A Steering Committee, consisting of the Chair and SecretaryGeneral of both sides and the T/C Chairs from both sides reviewed the Strategic

Plan and its progress in implementing the recommendations. The Panel has

implemented many of the recommendations of the Strategic Plan and will work

toward more complete implementation of the plan's recommendations.

The Panel approved the new Task Committee D (Wind Engineering) whose operating

charter was submitted at the 34th Joint Panel Meeting and met the criteria described in

the Strategic Plan.

The Panel endorses the following five proposed Task Committee Workshops or

Coordinating Committee Meetings during the coming year:

a. Task Committee (A), 5th U.S.-Japan Workshop on Geotechnical Earthquake

Engineering, Early 2003, San Diego, CA, USA.

b. Task Committee (B), 3rd U.S. -Japan Technical Coordinating Committee Meeting

on Auto-Adaptive Media (Smart Structural Systems), October 2002, Tsukuba,

Japan.

c. Task Committee (C), 3rd U.S.-Japan Workshop on Advanced Research on

Earthquake Engineering for Dams, June 2002, San Diego, CA, USA.

d. Task Committee (D), 3

rd U.S.-Japan Workshop on Design for Wind and Wind

Hazard Mitigation, Fall 2002, Seattle, WA, USA.

e. Task Committee (G), 18th U.S.-Japan Bridge Engineering Workshop, October

2002, USA.

In the event that T/C co-chairs consider it essential to conduct a joint meeting or workshop that is not included in this Resolution, the T/C co-chairs will make such a

request through their respective Secretary-General for approval by the U.S. and Japan

Joint Panel Chairmen.

The U.S. and Japan-side Task Committee Chairs, with concurrence of the Joint Panel

1
Chairmen, will develop and schedule their Workshops. Both sides' Secretaries-General

will be kept informed of the planning, conduct, and results of workshops or committee

meetings including resolutions and reports. Results of each activity conducted before the

35th Joint Panel Meeting will be presented at the 35th Joint Panel Meeting.

7. The Panel focused attention to the topic terrorism — the collapse of the World Trade

Center (WTC) and the Pentagon Wedge A by featuring a special Panel Meeting theme to

discuss this topic. Following the Panel Meeting, a technical site visit was scheduled to

visit the WTC site in New York City to discuss the structural collapses in greater detail.

8. The Panel recognizes the importance of continuing its joint research programs on Soil

Liquefaction and Countermeasures and on Auto-Adaptive Media (Smart Structural

Systems), and developing a U.S.-Japan coordinated research plan utilizing the NEES

facilities in the USA and diverse laboratories in Japan, including E-Defense.

9. The Panel resolved to explore collaboration on research in the area of progressive

collapse of buildings.

10. The Panel resolved to explore formation of a new Task Committee on Disaster

Information Technology to include information technology, public health, and security

aspects of development of next-generation building systems.

11. The U.S.- and Japan-sides will encourage the conduct of joint investigations following

earthquake and wind disasters in the U.S. and Japan.

12. The results of the Panel's work should be disseminated to improve the quality of life

globally. The Panel will make use of the internet to the extent possible to disseminate

information.

13. The Thirty-fifth Joint Panel Meeting of the UJNR Panel on Wind and Seismic Effects

will be held by the Japan-side Panel, in Japan in May 2003. The Japan-side Secretariat at PWRI will propose dates, program, location, and itinerary with the concurrence of the

U.S.-side Panel.

li

MANUSCRIPTS

SESSION 1

GEOTECHNICAL ENGINEERING

AND GROUND MOTION

Seismic Performance of Urban, Reclaimed and Port Areas

-Full Scale Experiment Using Blast Technique

by

Takahiro Sugano1 and Eiji Kohama2

ABSTRACT

A full scale lateral spreading experiment was

carried out on November 13

th

, 2001 at Tokachi

port in Hokkaido island, Japan. The primary

objective of the experiment was to assess the

performance of steel sheet pile quay walls

subjected to liquefaction/lateral spreading. Two

5.5 m deep quay walls were designed and

constructed, to investigate the damage mechanism

of steel sheet pile quay walls. They were

constructed with seismic design coefficients of

k=0.15 and k=0 separately. The test site was

reclaimed with dredged fine sand about 18

months before the test. The depth of the reclaimed

sand layer was about 8 m and the test field area was around 4800 m

2 (50 m by 96 m).

Controlled blast technique was used to induce

liquefaction/lateral spreading, and 127 blast holes

each consisting 4 kg explosive material for lower

part and 3 kg for upper part were charged. The

total weight of explosive material was 840 kg.

Various structures such as group pipe piles,

gas pipes and buried structures were installed

with measuring instruments. Two liquefaction

remediation methods and several new

measurement techniques were adopted. Forteen

organizations were participated in this project.

KEYWORDS: Full scale field experiment,

Liquefaction, Lateral spreading,

Reclaimed land

1. INTRODUCTION

The 1995 Hyogoken-Nambu earthquake

caused severe soil liquefaction in extensive areas of reclaimed land in Kobe. The soil liquefaction

induced large ground displacements in the

horizontal direction, which resulted in damage to

buried structure such as lifeline pipes and

foundations. Also, many caisson type quay walls

were damaged in Kobe Port during the

earthquake.

Generally, three sets of data can be used to

investigate the mechanism of damages;

1) Strong motion record nearby the damaged

facility,

2) Geotechecnical data, design conditions and

damage data ( i.e. deformation), and,

3) Numerical simulation or model test results

However, an important piece of information as

'behavior of structures during earthquake' is not

available. In this project, we focused on the

behavior of full scale structures during

liquefaction state and 14 organizations were

jointed in the project as shown in Table 1.

1 Head, Structural Dynamics Division, Geotechnical and Structural Engineering Department, Port and

Airport Research Institute, 3-1-1 Nagase, Yokosuka, Kanagawa 239-0826, JAPAN

2 Research Engineer, ditto

5
2. OUTLINES OF THE EXPERIMENT

The test site was located in Pier No.4 at Tokachi Port in Hokkaido. The Pier No.4 filling

construction was completed in June 2000 with use

of dredged fine sands. The experimental yard area was around 4800 m

2 (50 m by 96 m) and it was

separated into two parts. In one part, steel sheet

pile quay wall which was designed with seismic

coefficient of k=0.15 was installed the other part

the one designed with k=0 was employed. In this

paper, the k=0.15 part is refered as region (A),

and k=0 part as region (B). In region (B), 4%

slope (25 m width, 50 m long) were constructed

to induce lateral spreading.

The typical soil profile is shown in Fig.l. The

particle size distribution of the soils are shown in

Fig.2, as can be sheen most of the soil samples are

in 'possibility of liquefaction' zone.

2.1 Blast sequence

The blast hole is indicated in Fig.3 from Al to A54 , from Bl to B48 and from CI to C25. The

first ignitions were Al and Bl at the same time.

Then, ignite next holes in order with 500 ms

time interval which showed snake walking in

plan view. Each hole consisting 4 kg explosive

material for lower part and 3 kg for upper part

were charged with 200 ms time interval, the

ignite order was from lower part to top part.

2.2 Lift up mechanism (Fig.3)

In region (A), The Japanese Geotechinical

Society installed eight buried structures such as sewage water pipes to investigate the lift up

mechanism during liquefaction and the

countermeasures. The dimensions of the installed

cylindrical structures were 1 m in diameter and

3.8 m in length.

2.3 Performance of Cement deep mixing soil

improvement/Permeable grouting improvement

during liquefaction

UC Berkeley and Japan Cement Deep Mixing

Association installed four cement deep mixed

specimens in region (A) as a 3.4 m by 3.4 m foot

print cement deep mixing improvement to depth

of 8 m (full improvement of reclaimed layer), 5.5

m , 3.5 m and no treatment.

Permeable Grouting Method Association and

Port and Airport Research Institute tried to

injected 8 hrs gel time chemical grout in between

sheet pile quay wall and the anchor pile to

investigate the efficiency of permeable grouting

method as shown left bottom zone of region (A)

in Fig.3.

2.4 Performance of steel sheet pile quay walls

Port and Airport Research Institute, Japan

Dredging and Reclamation Engineering

Association and Japan Steel Pipe Pile Association

constructed two steel sheet pile quay walls with

different design seismic coefficients including

strain gauges and load cells. In region (A), the

quay wall with seismic coefficient of k=0.15 ,and

in region (B), the quay wall with k=0 were

constructed.

At the tie rod connectors just behind the sheet

piles in region (B), special explosives were used

so as to cut the tie rod to induce lateral spreading

during liquefaction.

2.5 Performance of lifelines subjected to lateral

spreading

UC San Diego and Waseda University carried

out the pile experiment. The three sets of pile

foundations were installed behind the k=0 quay

walls to determine the effect of pile groups on the

pile response.
300 mm diameter steel pipe piles in pile

groups having 900 mm spacing and a concrete

pile cap (three single piles, 4-pile group, and

9-pile group) were constructed.

Two pipes were installed to intersect the

lateral spreading direction at 90 degree angle, and

one pipe was installed in the lateral spreading

direction as shown in the right bottom part of

Fig.3.

Ten wireless transmition dynamic GPS

devices (10 Hz time interval) were installed and

attached to the structures to measure the

deformation of structures during lateral spreading.

2.6 Investigation on the mechanism of lateral

spreading

Waseda university installed 11 GPS devices

on the surface of slope with 1 Hz measurement

interval, and installed vertical optical fiber

modules to measure the deformation inside the

slope.

2.7 Evaluation of impervious sheet properties

when large deformations occur during and after

earthquake

National Institute for Land and Infrastructure

Management Yokosuka, installed the impervious

sheet used for garbage dumping ground. Many of

Japanese garbage dumping grounds are located in

coastal area. To keep the garbage material and

polluting matter inside the designated area, the

performance of impervious sheet is important.

2.8 Survey and Sounding techniques

Port and Airport Research Institute conducted

several sounding tests as follows: SPT-blow count,

seismic cone penetration test, portable cone

penetration test, automatic Swedish weight

sounding, and dilatometer test. These tests were

conducted about a couple of months before, the

day before and after the blast. It is also planed to

perform them in June 2002.

National Institute of Advanced Industrial

Science and Technology research team used the

resistivity method to measure the density profile

of the liquefied soil.

Earthquake Research Institute, Tokyo Univ.

and Port and Airport Research Institute compared

the data obtained from the dynamic GPS installed

in region (A) and conventional laser survey data.

Chuo Univ. used the radio controlled

helicopter to take photos of the experimental site

before and after the blast and automatic computer

processing technique to measure the deformation.

Chuo Univ. also installed CCD camera into the

vertical transparent cylinder to visualize the soil

profile during the test.

Waseda Univ. team tried to detect the

liquefied soil layer profile during liquefaction by

using acoustic tomography with pseudo random

binary sequence code technique.

UC Berkely and Port and Airport Research

Institute installed both conventional strain gauge

type accelerometers and wireless accelerometers

which using the micro electro mechanical systems

and bluetooth transmission. In this project around

800 sensors were installed and total length of

signal cable between sensor and amplifier was

about 40 km. The data obtained from wireless

accelerometers showed good agreement with the

data obtained from conventional accelerometers.

2.9 Re-liquefaction phenomena

Civil Engineering Research Institute of

Hokkaido is interested in the re-liquefaction

phenomena and curried out the post experiment

on Dec. 14

th

, 2002 with UC San Diego and Port

and Airport Research Institute. In region (B), the

7
controlled blast were applied with the group pile

foundations as its center.

3. SUMMARY OF RESULTS

The overall deformation of the test field is shown in Fig.4. The maximum horizontal

displacement of quay wall in region (A) was

about 30 cm, region (B) was about 130 cm

respectively. Due to the horizontal movement of

quay walls, lateral spreading was occurred in

region (B). However, three sets of group piles

played a role in control works against lateral

spreading as shown in Fig.5.

In this experiment, acceleration caused by

controlled blasting could not simulate actual

earthquake acceleration as shown in Fig.5. Fig.5

shows an acceleration time history recorded at

region (A) quay wall, the maximum acceleration

was over 10 g. On the other hand, excess pore

water pressure was built up overburden pressure

as shown in Fig.6.

The bending moment distribution of initial

condition, maximum displacement condition and

residual condition are shown in Fig.7. The failure

of sheet pile quay walls occurred due to the

reduction of resistance of anchor pile during

liquefaction, then sand water mixture fluid push

the quay walls.

Many kinds of challenges to investigate

seismic performance of urban, reclaimed and port

were conducted in collaboration with 14

organizations. The reports in this paper include

wide filed subjects related to liquefaction/lateral

spreading phenomena. The analysis works is

presently continuing by individual organization

participated in this experiment. The results will be

presented in the future.

8
Table 1 Organizations of the research project

Organization Research content Role

Port and Airport Research

Institute (PARI)

1. Investigation on the dynamic characteristics of ground before and after

liquefaction

2. Investigation on seismic behavior of sheet piles, anchor piles and tie rods/wires

supervision of entire experiment, measurement of behavior of sheet piles and

analysis

National Institute for Land and

Infrastructure Management

(NILIM)

1. Evaluation of impervious sheet properties when large deformations occur

during and after earthquake 2. Investigation on characteristic of microtremor

before and after earthquake

analysis of experimental results

Japan Dredging and

Reclamation Engineering

Association (JDREA)

Investigation on seismic behavior of sheet piles, anchor piles, tie rods and

surrounding ground

execution management and analysis

Japanese Association for Steel Pipe Piles (JASPP)

Investigation on seismic behavior of sheet piles, anchor piles, tie rods and

surrounding ground

measurement of sheet piles displacement and

analysis

University of California, San

Diego (UCSD)

Investigation on behavior of pile groups and buried pipes

Measurement and analysis about interaction between structure and ground during lateral flow Cement Deep Mixing

Association (CDM Association)

Investigation on properties of ground improved by cement deep mixing method

cement mix design, execution management

and analysis

University of California, Berkrey (UCB)

Investigation on properties of ground improved by cement deep mixing methodmeasurement of improved ground behavior and analysis The Japanese Geotechnical

Society (JGS)

Investigation on behavior of buried pipes imposed to up-lift pressure of

liquefaction

measurement of underground pipe behavior and analysis

Waseda University (Waseda Univ.)

1. Investigation on ground movement during and after lateral flow and

behavior of buried single pile 2. Investigation on elastic wave velocity in liquefied ground 3. Measurement of vibration at adjacent buildings 4. Measurement of underground displacements

acoustic tomography, measurement of underground displacement by using optical

fiber, GPS, etc.

Permeable Grouting Method

Association (PGM Association)

Confirmation of efficiency of permeable grouting method execution management and analysis

Tokyo University, Earthquake Research Institute (Tokyo Univ.)

Measurement of 3D dynamic displacements by GPS measurement by dynamic GPS

\_nuu (jiuvcrMiy (Chuo Univ.)

1 Invpcfi nation on u/ntpr film nhf*nnmpnnn nrpnrrpH hv linnpfaption

2. Measurement of ground displacements

mpaQiirpmpnt hv HArphnlp-famprn anH

radio-control helicopter National Institute of Advanced

Industrial Science and

Technology (NIAIST)

Investigation on underground density profile of liquefied soil measurement by resistivity method

Civil Engineering Research

Institute of Hokkaido (CERI)

Investigation on behavior of re-liquefied ground analysis of experimental results

9
10
*

11
12

14
15
16
17
18
19

A Review of Empirical Evidence For Site Coefficients in Building-Code Provisions

by

Roger D. Borcherdt a)

ABSTRACT

F F

Site-response coefficients, a and v

, used in

U.S. building code provisions are based on

empirical data for motions up to O.lg. For larger

motions they are based on theoretical and

laboratory results. The Northridge earthquake of

17 January 1994 and other recent earthquakes

have provided significant new sets of empirical

data up to 0.5g. These data together with recent

site characterizations based on shear-wave

velocity measurements provide empirical

estimates of the site coefficients at base

accelerations up to 0.5g for Site Classes C and

F F

D. These empirical estimates of a and v as well as their decrease with increasing base

acceleration level are consistent at the 95

percent confidence level with those in present

building code provisions, with the exception of

F

estimates for a at levels of 0.1 and 0.2 g, which are less than the than the lower

confidence bound by amounts up to 13 percent.

The site-coefficient estimates are consistent at the 95 percent confidence level with those of

several other investigators for base accelerations

greater than 0.3 g. These consistencies and

present code procedures indicate that changes in

the site coefficients are not warranted.

KEYWORDS: Amplification, Building Codes,

Design Spectra, Ground-motion, International

Building Code, NEHRP Provisions, Northridge

Earthquake, Shear-Wave Velocity, Site

Coefficients, Site Classes, Uniform Building

Code

1.0 INTRODUCTION

Present US building codes are based on

empirical estimates of site coefficients, Fa and

Fv

, derived from recordings of the Loma Prieta

earthquake at the O.lg level (Borcherdt 1993,

1994) and derived at higher levels using

numerical modeling and laboratory results

(Seed, et al., 1994). The Northridge earthquake

of 17 January 1994 and other recent earthquakes

have provided sets of ground-motion recordings

with peak ground accelerations at levels up to 0.5g. These new data provide an opportunity to further evaluate site coefficients as currently

specified in US building codes.

This paper provides a review of the empirical

evidence for site coefficients as derived by

analyses of Northridge earthquake recordings

(Borcherdt, 2002) and those derived by a number of other investigators using other

databases and procedures.

2.0 COMPARISON OF SITE COEFFICIENTS

DERIVED USING VARIOUS METHODS

Estimates of the short-period Fa and mid-period

Fv site coefficients have been derived relatively

recently by a number of investigators using a

variety of databases and procedures. Results

of these investigations are compared herein with

those derived from Northridge by Borcherdt

(2002). Estimates of Fa and Fv together with

corresponding 95 percent confidence intervals

for each site class as derived by Borcherdt

(2002) are plotted in Figures 1 and 2. Corresponding code values

(F ,FV ) and results derived by

code Table code Table Crouse and McGuire (1996), Dobry et al. (1999), Joyner and Boore (2000), RodriguezMarek et al. (1999), Silva et al. (2000), and

Stewart et al. (2001, 2002) are provided for

comparison.

Crouse and McGuire (1996) derived estimates

of site coefficients using a set of empirically

based attenuation curves based on strongmotion recordings from a select group of 16

U.S. earthquakes, which included the Landers

earthquake, but was completed before the

Northridge data could be incorporated. Site

a) United States Geological Survey

Menlo Park, CA 94025

21
classifications were based on a separately

compiled database. The site coefficients were

derived from empirical predictions at various

magnitude and acceleration levels. Ratios of

spectral levels at specific periods were

computed from averages for the magnitude

acceleration pairs, from which averages for the

short and mid period band were computed.

Dobry et al. (1999) derived amplification ratios

using the Northridge strong-motion data from

ratios of response spectral ordinates for nearby

soil-rock pairs. Their database is a subset of that

used by Borcherdt (2002). They derived

estimates from ratios computed using a hypocentral distance norm (method 1) and from

ratios computed by normalizing the response

spectra for various soil sites to the

corresponding value predicted by Silva using

the Abrahamson and Silva (1997) attenuation

relation for rock (method 2). Estimates of the

short and mid-period site coefficients were

derived from averages of the response spectral

ratios over the short- (0.1 - 0.5 s) and mid- (0.4 - 2.0 s) period bands. Estimates of the mean

and standard deviation were derived for Site

Classes C and D from a subset of sites for which

the ratios were considered most reliable.

Averages of their ratios for base acceleration

intervals of 0.04 - 0.14 g, 0.15 - 0.24 g, and 0.25

- 0.35 g are plotted (Figures 1 and 2).

Joyner and Boore (2000) derived estimates of

Fa and Fv by adding a new term to their

regression relation of the form

(a6 +a7 PSVref )\og(V/Vref ),

(3)

where V is the average V30 for the site class, Vref is the average V30 for the reference site

condition, PSVref is the predicted pseudo

spectral velocity for the reference site condition, a6 and a7 are coefficients determined by

regression, and distance is measured as closest

distance to projected rupture surface. Their

estimates are based on a database, which does

not include the Northridge, Landers, or Loma

Prieta data sets. Their estimates are not

averaged over a period band, but correspond to

spectral ratios at 0.2 and 1.0 seconds. Their

estimates of Fa and Fv as derived with respect

to a reference site velocity of 1068 m/s are plotted (Figures 1 and 2).

Rodriguez-Marek et al. (1999) used the strongmotion recordings from the Loma Prieta and

Northridge earthquakes. They classified the sites

based on general geologic and geotechnical

information, but did not consider shear-velocity

measurements essential to their classification

scheme. They developed attenuation relations

for each earthquake and site class from which

they developed amplification factors with

respect to an attenuation relation for site class B.

Silva et al. (2000) developed generic shearwave velocity profiles for the surficial geologic

units in California. Silva used these profiles and

a random vibration theory equivalent linear

model to estimate amplification factors as a

function of frequency. He developed estimates

based on both Peninsular Ranges and EPRI

models of randomized material profiles. He also

developed curves for conditions considered

most appropriate for both the western and

eastern U.S. Estimates inferred from figures in

Silva et al. (2000) for the Peninsular Ranges

model are plotted (Figures 1 and 2).

Stewart et al. (2001) used a large database

recently compiled by Silva (PEER,

http://peer.berkeley.edu/smcat/). It includes

recordings from the Northridge, and other

earthquakes that have occurred up until 1999.

Recording sites were classified using mapped

surface geology. Their estimates of Fa and Fv were derived from ratios of response spectral

acceleration as computed from the recordings at the site and a reference peak acceleration as predicted by the attenuation functions of

Abrahamson and Silva, 1997. Their estimates

of Fa and Fv are derived from averages over

the appropriate period band. They used linear

regressions of the logarithm of estimates for Fa and Fv on the logarithm of predicted reference

peak ground acceleration to derive the estimates

of the amplification factors as a function of

input acceleration level.

The procedures and database used by Borcherdt

(2002) differs in two important respects from

those of other investigators. The amplification

factors are computed with respect to that for a nearby reference site to minimize effects of

variations in source radiation pattern and crustal

22
propagation path. The estimates are based only

on the data recorded from the Northridge

earthquake. The sites are classified using shear

velocity as either measured or estimated for

each recording site. Classification of the sites

using shear velocity is preferred because seismic

response correlates better with shear velocity

than with geologic age. In addition, shear

velocity is preferred because the actual site

characteristics at several recording sites were

found to differ from those shown on the

geologic maps. The reason for this difference is

that several of the sites are located near a

geologic boundary that may have been

generalized or the geology mapped at the

surface is not the same as that within a few feet

of the surface.

The estimates of site coefficients by various

investigators (Figures 1 and 2) vary depending

on the database, the reference ground motion,

the site-classification method, and the procedure

used to infer the resultant site factors. In

addition to previously stated differences in

procedures to develop reference motions and

resultant estimates of the site coefficients, the

site classification method is thought to be a

significant contributor to uncertainty in the

various estimates. Limited shear-wave velocity

data at most of the strong-motion recording sites

necessitates the use of mapped geology to classify the sites. Unfortunately, units on most

geologic maps have been mapped for purposes

of inferring geologic history and not seismic

response. For example, in the San Francisco

Bay region (Borcherdt, 1994) soil units mapped

as quaternary alluvium (Qal) excluding the

Holocene units may contain deposits ranging

from fine-grained clays and sands with vs near 250 m/s to over consolidated, very dense

course-grained sandy gravels with v near 600

m/s. Similarly, rock units mapped as Tertiary in

age may vary from firm to soft sandstones and

siltstones with v near 300 m/s to hard to firm

rhyolites with little or no weathering and vSjq near 1000 m/s. Without shear-velocity

information, sites with rocks of Tertiary and

Mesozoic age tend to be classified as class B

sites resulting in higher average reference

motions and hence lower estimates for the site

coefficients than would be obtained were sites

correctly classified using shear-velocity

information. The uncertainties induced by

geologic classifications are expected to be a

significant contributor to the differences in the

estimates derived by the various investigators

using the different databases from different

geologic and tectonic regions. As a further

explanation of some of the uncertainties in

estimates derived by various investigators,

results not based on Northridge strong-motion

data are necessarily derived from data sets with

fewer data with base acceleration levels above

0.2g, and hence, would not be expected to show

as large a dependencies on peak acceleration. In

addition, estimates derived at individual periods

of 0.2 and 1.0 seconds are expected to be less

and greater than those, respectively, computed

as averages over the short (0.1-0.5 s) and midperiod (0.4-2.0 s) bands.

Results derived by Stewart et al. (2001) are consistently less than those derived by other

investigators and only within the 95 percent

confidence band for peak accelerations greater

than about 0.35 g to 0.42 g depending on site

class. Stewart et al. (2001) attributes this

difference to the fact that their estimates are referenced to "soft rock" as opposed to "firm to hard" rock used for the code, however

uncertainties caused by the necessity to classify

sites using geologic age may also be a

contributor. Results of Stewart (2002)

referenced to a velocity of 1050 m/s are also

shown. These estimates compare better with

those in the code. Further discussion will be

limited to the estimates derived from averages

by Crouse and McGuire (1996), Dobry et al. (1999), Rodriguez-Marek et al. (1999), Silva et

al. (2000) for the Peninsular Ranges, and

Stewart (2002).

Comparison of the short-period Fa estimates

derived from averages for Site Classes C and D

shows that the estimates of Silva et al. (2000)

for the Peninsular Ranges and Rodriguez-Marek

et al. (1999) are near or within the 95 percent

confidence band for the sample estimates of the

ordinates to the true population regression line

derived by Borcherdt (2002) for each base

acceleration level. Their estimates and those of

Crouse and McGuire (1996) and Borcherdt

(2002) exceed the code values for each base

acceleration level. Estimates by Crouse and

McGuire (1996) and Stewart (2002) are below

the 95 percent confidence bound for base

23
acceleration levels near 0.2 g and less. The

value of Dobry et al. (1999) for Site Class D at the 0.1 g is significantly less than the code

value. Their values for Site Class C at O.lg and

0.3 g are within the 95 percent confidence band,

but their value at 0.2 g is not. With the

exception of the estimates by Crouse and

McGuire (1996) for Site Class C estimates

derived by each investigator decrease with

increasing base acceleration.

Comparison of the mid-period Fv estimates for

Site Class C shows that the estimates of Crouse

and McGuire (1996) and Silva et al. (2000)

exceed the code values and are within the

confidence band derived by Borcherdt (2002)

for each base acceleration level. The estimates

of Rodriguez-Marek et al. (1999) are

significantly less than the code values and

outside the 95 percent confidence band except

for base acceleration of 0.4g. The Dobry et al. (1999) values do not show a well-defined

dependence on peak acceleration and are within

the confidence band only at 0.3 g. Estimates

derived for Site Class C by Crouse and McGuire

(1996) and Silva et al. (2000) do not show a dependence on base acceleration.

Comparison of the mid-period Fv estimates for

Site Class D shows that the estimates of Silva et

al. (2000) exceed the code values and are within

the confidence band for each base acceleration

level. Estimates derived by Crouse and McGuire

(1996) and Rodriguez-Marek et al. (1999) are within the confidence band for base

accelerations 0.3 g and greater, but below the

band for accelerations 0.2g and less. Each of

these estimates shows a dependence on base

acceleration.

3.0 CONCLUSIONS

Recent estimates of site coefficients derived by

various investigators vary depending on the

database, the reference ground motion, the siteclassification method, and the procedure used to infer the resultant site factors. These variations

tend to suggest the lack of a consensus for

modifying the code factors. However, the

consistent tendency for the estimates of Fa derived by Crouse and McGuire (1996), and

Rodriguez-Marek (1999), Silva et al. (2000),

and those herein to exceed the code values at each base acceleration suggests that some

increase in the Fa factors might be appropriate

for Site Classes C and D. Similarly, the

tendency for the estimates of Fv by Crouse and

McGuire (1996), Silva et al. (2000), and

Borcherdt (2002) to exceed those of the code for

Site Class C might be used to argue that some

increase in Fv for Site Class C is appropriate.

An increase in estimates of Fv for Site Class D

can be argued on the basis of the consistency

between the Silva et al. (2000) results, the

results of Borcherdt (2002), and the Joyner and

Boore result at 1 second, but not the results of

the other investigators. In summary, variation in

the estimates derived by various investigators

and the uncertainly associated with the various

estimates does not support a significant change

in the site coefficients Fa and Fv as currently

specified in US building codes.

4.0 ACKNOWLEDGMENTS

Discussions during the initial research with two

special colleagues, the late W. B. Joyner and the

late H.P. Liu, were helpful.

5.0 REFERENCES

Borcherdt, R. D., 1994, Estimates of sitedependent response spectra for design

(Methodology and Justification), Earthquake

Spectra, 10, 617-653.

Borcherdt, R. D., 2002, Empirical evidence for

site coefficients in building-code provisions,

Earthquake Spectra, accepted.

Crouse, C. B., and McGuire, J. W., 1996, Site

response studies for purpose of revising NEHRP

seismic provisions, Earthquake Spectra, 12,

407-439.

Dobry, R., Martin, G. M., Parra, E., and

Bhattacharya, A., 1994, Development of sitedependent ratios of elastic response spectra

(RRS) and site categories for building seismic

codes, Proceedings NCEER, SEAOC, BSSC

workshop on site response during earthquakes

and seismic code provisions, University of

Southern California, Los Angeles, California,

November 18 -20, 1992.

24
Dobry, R., Ramos, R., and Power, M. S., 1999,

Site factors and site categories in seismic codes,

Technical Report MCEER-99-0010, 81 pp.

International Council of Building Officials

(ICBO), 2000, 2000 International Building

Code, http://www.icbo.org/.

NEHRP (National Earthquake Hazards

Reduction Program) Recommended Provisions

for Seismic Regulations for New Buildings,

1994 Edition, 1995, Federal Emergency

Management Agency Reports FEMA 222A,

FEMA 223A.

NEHRP (National Earthquake Hazards

Reduction Program) Recommended Provisions

for Seismic Regulations for New Buildings,

1997 Edition, 1998, Federal Emergency

Management Agency, Washington, D. C,

FEMA 302, FEMA 303.

Rodriguez-Marek, A., Bray, J. D., and

Abrahamson, N., 1999, Characterization of site

response general categories, PEER Report

1999/03, Pacific Earthquake Engineering

Research Center, Berkeley, Ca.

Seed, R. B., Dickenson, S. E., and Mok, C. M.,

1994, Site effects on strong shaking and seismic

risk; recent developments for seismic design

codes and practice, American Society Civil

Engineering Structures Congress, 12, 573-578.

Silva, W., Darragh, R., Gregor, N., Martin, G.,

Abrahamson, N., and Kircher, C, 2000,

Reassessment of site coefficients and near-fault

factors for building code provisions, USGS

EHRPprogram report 98-HQ-GR-1010.

Stewart, J. P., Liu, A. H., and Baturay, M. B.,

2001, Amplification factors for spectral

acceleration and duration in active regions,

California Strong-motion Instrumentation

Program Report 2001- , in press.

25
C

o

o

I

o

4>a

o

.c

Site Class D, Fa

Fs S0.3 = - 0.18 Base Accel. + 2.23

95% Conf. Limit for Ordinate

Fa code

o Stewart eta I. (2001)

Crouse & McGuire (1996, Table 6)

a Joyner & Boore (2000, Vs ref. = 1068m/s)

Dobry, et a I., (1998, Method 2)

Rodriguez-Marek, et al., (1999)

Silva etal. (2000, Penn. Range, WUS)

Stewart, 2002

1 2 3 4 5

Base Acceleration (m/s2) = Surface Acceleration / Acceleration Ratio

(0 u. 4

c

o

n

o

SS 3

a

I

o

.2 2 »a

r

o

5 1

Site Class C, Fa Fs SO. 3 = -0.08 Base Accel. + 1 95% Conf. Limit for Ordinate

Fa code

71

o Stewart etal. (2001)

Crouse & McGuire (1996, Table 6)

Dobry, et al., (1998, Method 2)

a Joyner & Boore (2000, Vs ref. = 1068m/s)

Rodriguez-Marek, et al., (1999)

Silva et al. [2000, Penn. Range, WUS)

• Stewart, 2002

Fa code ~ -0.06 Base accel. + 1.28

%

0 1 2 3 4 5

Base Acceleration (m/s2) = Surface Acceleration / Acceleration Ratio

Figure 1. Empirical estimates of the short-period site coefficient Fa for site classes D and C as

derived by several investigators.

26
>

LLC

o

* 5

o

is

I

o

o

a

Site Class D, Fv Fm S1.0 = -0.20 Base Accel. + 2.82

95% Conf. Limit for Ordinate

Fv code

o Stewart eta I. (2001) a Crouse & McGuire (1996, Table 6)

Dobry, et al., (1998, Method 2)

a Joyner & Boore (2000, Vs ref. = 1068m/s)

Rodriguez-Ma re k, et al., (1999)

Silva et al. (2000, Penn. Range, WUS)

Stewart, 2002

Fv code ~ -0.22 Base Accel. + 2.52

0 1 2 3 4 5

Base Acceleration (m/s2) = Surface Acceleration / Acceleration Ratio

6

I

6

I 5

n

o

1*

o 3

o

"C0

Site Class C, Fv Fm S1.0 = -0.10 Base Accel. + 2.05

95% Conf. Limit for Ordinate

Fv code

o Stewart eta I. (2001)

Crouse & McGuire (1996, Table 6)

Dobry, et al., (1998, Method 2)

a Joyner & Boore (2000, Vs ref. = 1068m/s)

Rodriguez-Marek, et al., (1999)

Silva et al. [2000, Penn. Range, WUS)

Stewart, 2002

Fv code - -0.10 Base accel. + 1.8

1 2 3 4 5

Base Acceleration (m/s2) = Surface Acceleration / Acceleration Ratio

Figure 2. Empirical estimates of the mid-period site coefficient Fv for site classes D and C as derived

by several investigators.

27

A Practical Procedure for Formulating Level 2 Earthquake

Motions Based on Scenario Earthquakes

by

Shojiro Kataoka1 , Jun Murakoshi2 and Keiichi Tamura3

ABSTRACT

A procedure for formulating Level 2 earthquake

motion (L2 motion) is presented. Tokyo Bay

mouth area is given as an example of target site. The most influential earthquake out of possible

events around the area is selected as a scenario

earthquake; its location and size are assumed the

same as those of the 1923 Kanto earthquake.

The L2 motion is evaluated following numerous

strong motion simulations assuming various

source fault models for the scenario earthquake.

A stochastic Green's function method is proposed

and applied to the simulations. The L2 motion is compared with probabilistic seismic hazard and

the input motions specified in the current design

specification for highway bridges.

KEY WORDS:Level 2 Earthquake Motion

Scenario Earthquake

Strong Motion Simulation

Stochastic Green's Function

1. INTRODUCTION

The design specifications for highway bridges,

railroad bridges, port facilities, and so on, were

revised in Japan following the 1995 Hyogo-ken

Nanbu earthquake, which caused great damage to

infrastructure including those structures. As for

highway bridges, the main features of the revision

regarding seismic design force is that Type U

ground motion, i.e. ground motion from inland

direct strike type earthquake like the Hyogo-ken

Nanbu earthquake, is to be taken into account

besides Type I ground motion, i.e. ground motion

from plate boundary type large-scale earthquake.

These two types of ground motions are both

classified as those with high intensity and low

probability to occur during the service period of

the bridge. They are applied to seismic design

based on ductility design and dynamic analysis

methods. The target seismic performances of

bridge for them are set to prevent fatal damage for

bridge of standard importance and to limit damage

for bridge of high importance. These design

ground motions with high intensity are called

"Level 2 earthquake motions" in the current

design specification for highway bridges, which

has been newly revised in 2002. On the other

hand, design ground motions that cover ground

motion highly probable to occur during service

period of bridge are called "Level 1 ground

motions". The target seismic performance of

bridge for Level 1 earthquake motions is usually

1 Senior Researcher, Earthquake Disaster Prevention Division, Research Center for Disaster Risk

Management, National Institute for Land and Infrastructure Management, Ministry of Land,

Infrastructure and Transport, 1 Asahi, Tsukuba 305-0804, Japan

2 Team Leader, Bridge Structure Research Team, Structure Research Group, Independent Administrative

Institution Public Works Research Institute, 1-6 Minamihara, Tsukuba 305-8516, Japan

3 Team Leader, Ground Vibration Research Team, Earthquake Disaster Prevention Research Group,

ditto.

29
set to have no damage.

The new design specification for highway bridges

(Japan Road Association, 2002), DS02 from now

on, describes Level 1 and 2 earthquake motions as

the standard acceleration response spectra, as well

as the previous edition published in 1996 (DS96).

DS96 had requested designers to use these design

ground motions without exception. DS02,

however, doesn't. The designers need to set up

Level 1 and 2 earthquake motions in case that

they can do so appropriately with adequate

consideration regarding seismicity, ground

structure and soil condition around the target site

in addition to strong motion records etc.

Attenuation relations, which are derived by

regression analyses of strong motion records, have

been used to estimate peak amplitude or response

spectrum from earthquake magnitude and distance

between source and site. Since those records are

mostly moderate and naturally lacking in very

high intensity, the attenuation relations may have

insufficient accuracy to estimate earthquake

motion with very high intensity. Thus, it is

adequate to use them to set up Level 1 earthquake

motions with consideration of regional difference.

On the other hand, we need to utilize the other

kind of methods to estimate high-intensity

earthquake motion at the target site with sufficient

accuracy in order to set up Level 2 earthquake

motions.

We have been studying and developing a practical

procedure to estimate strong ground motion and

set up Level 2 earthquake motion. The

procedure consists of selection of scenario

earthquakes for the target site, simulation of

strong ground motions caused by the scenario

earthquakes with consideration of uncertainty of

source parameters, and formulation of Level 2

earthquake motion. In this paper, the procedure

is presented giving Tokyo Bay mouth area as an example of target site.

2. SELECTION OF SCENARIO

EARTHQUAKES

Earthquake motion possible to be observed at a

site is different from that at the other site. It is not only because of the difference of ground

structure and soil condition but earthquakes

possibly occur around each site. Therefore, it is

desirable that Level 2 earthquake motions for each

site are determined with consideration of a set of

scenario earthquakes that are selected based on

the tectonic setting of the region (Japan Society of

Civil Engineers, 1996).

In this paper, Kannonzaki site (35.250N,

139.750E) in Tokyo Bay mouth area is given as an example of target site to evaluate Level 2

earthquake motion and a scenario earthquake is

selected based on the occurrence of past

earthquakes and distribution of active faults

around the site. Figure 1 shows epicenters of the

past earthquakes (JMA magnitude Mj >=6.5)

occurred around the site and distribution of active

faults. The 1923 Kanto earthquake (Mj = 7.9)

has the largest magnitude among the past

earthquakes and its fault plane spreads beneath the

site (see Figure 3). The largest magnitude

possibly caused by the active faults in Figure 1 is Mj = 7.3 (Matsuda, 1995). Thus, recurrence of

the 1923 Kanto earthquake can be considered as

the most influential one and is selected as the

scenario earthquake for Level 2 earthquake

motion for Kannonzaki site.

3. SIMULATION METHOD

A stochastic Green's function method is proposed

here to simulate strong motion caused by scenario

earthquakes. As shown in Figure 2, the fault

30
plane is divided into Nt x Nw subfaults to take the

fault rupture process into account. Then the

ground motions due to the subfaults are calculated

and summed up to synthesize the ground motion

due to the target source fault (Irikura, 1986).

The ground motion from each subfault is also the

sum of ground motions from hypothetical element

earthquakes of which location, length, and width

are the same as the subfault but the slip is l/Nd. Then the seismic moment, which indicates the

total energy released by an earthquake, of the

element earthquakes should be set to l/( Nt x Nwx Nd) of the scenario earthquake. The ground

motion due to the element earthquakes can be

produced to fit observed ground motions in a

stochastic sense and used as Green's function

(stochastic Green's function). In this paper, the

Green's functions are generated by fitting their

acceleration response spectra and envelopes to

empirical models of those of ground motion on

engineering bedrock of which S-wave velocity is

700[m/s] (Satoh et al., 2002). Thus, the resulting

ground motion due to the target event is that on

the engineering bedrock.

4. STRONG MOTION SIMULATION

4.1 Source fault models of the scenario

earthquakes

The 1923 Kanto earthquake is considered

recurrent but not with the same rupture process.

Thus, it is necessary to consider uncertainty of

source parameters to be prepared for future

earthquakes. The locations of rupture starting

point and asperities and rupture velocity are chosen here as uncertain parameters and their

effects on strong motion are examined by

changing them systematically, while the fault

plane geometry is fixed.

Wald and Somerville (1995) proposed a

variable-slip fault rupture model of the 1923

Kanto earthquake as shown in Figure 3. The

fault model has the length of 130[km], width of

70[km], strike of 290[deg], dip of 25[deg]. The

rupture starts from the hypocenter and propagates

radially at rupture velocity of 3.0[km/s]. The

epicenters of the 1923 Kanto earthquake in

Figures 1 and 3 are located differently because the

epicenter in Figure 3 was determined

independently of JMA using local and teleseismic

P-wave arrival times.

Kanamori (1991) proposed an average empirical

relationship between seismic moment, Mo[N m],

and rupturearea, S[km2

], as Mo = 1.23 x 1015 S

3/2

assuming stress drop is 3.0[MPa]. Substituting

5=130 x 70=9100[km2

] into the equation yields

Mo = 1.06 x 10

21 [N m], which is more than 30%

larger than that of the 1923 Kanto earthquake,

7.8xl020[N m], estimated by Wald and Somerville

(1995). Here we set the seismic moment of the

scenario earthquakes to 1.06 x 10

2, [Nm] as it is more likely to occur in the future.

Somerville et al. (1999) proposed a procedure to

identify rectangle asperities from a variable-slip

model and applied it to 15 crustal earthquakes.

Figure 3 shows the rectangle asperities identified

from the variable-slip model of the 1923 Kanto

earthquake by the procedure. Since two

asperities have been identified here, we call the

larger one "Asp#l" and the smaller "Asp#2".

The average slip over Asp#l and Asp#2 become

2.00 times and 2.09 times of the average slip over

the entire fault. Those values are very close to

the value 2.01, which Somerville et al. (1999)

found as the average of asperities of the 15 crustal

earthquakes. Asp#l and Asp#2 cover 17% and

8.5% of the area of the entire fault, respectively.

Based on these characteristics of the source fault

of the 1923 Kanto earthquake, we assume: the

31
source faults of the scenario earthquakes have one

or two asperities; in case of one asperity, it covers about 25% of the area of the entire fault; in case

of two asperities, the larger and the smaller cover

about 17% and 8.5% of the area of the entire fault,

respectively; the average slip over the asperities

are just 2 times of the average slip over the entire

fault, i.e. 3.03[m]. The background, the rest of

the entire fault, is set to have uniform slip of

0.91 [m] to maintain the average slip over the

entire fault.

As shown in Figure 4, the locations of asperities

and rupture starting point are changed to examine

effects on strong motion. The rupture velocity is

also set to have three different values: 2.6, 3.0,

and 3.3[km/s]. Now the source fault models of

the scenario earthquakes have 9 patterns of the

locations of asperities, 5 locations of rupture

starting point, and 3 rupture velocities so that we

have total of 9x5x3 = 1 35 source fault models

for the following strong motion simulation.

4.2 Variation of ground motion due to uncertainty

of source parameters

Strong motion simulations are carried out using

the 135 source fault models by the stochastic

Green's function method explained earlier in this

paper. Figure 5 shows the acceleration response

spectra of the 135 simulated strong motions. We

can see and understand that the acceleration

response varies about 5 times due to the variety of

the source fault models and it is very important to conduct such simulations to set up design ground

motion for prevention of unexpected damage. It

is also shown that the simulated strong motions can be much stronger than the standard

acceleration response spectrum of the Type I ground motion specified in DS02.

5. EXMAMPLE OF LEVEL 2 EARTHQUAKE

MOTION

The acceleration response spectra of the 135

simulated strong motions are compared with those

with 2, 5, and 10 % probability of exceedance in

100 years derived by probabilistic seismic hazard

analyses (Nakao et al., 2002). An attenuation

model of acceleration response spectrum

(Takahashi et al., 2000) is employed for the

analyses. These three curves are within the

variation range of the 135 acceleration response

spectra.

If, for example, the Level 2 earthquake motion is

set to have the acceleration response spectrum

consists of three straight lines shown in Figure 6,

the waveforms of the Level 2 earthquake motion

can be as shown in Figure 7. One of the

simulated waves is chosen and its Fourier

amplitude spectrum is modified so that its

acceleration response spectrum fits to the target

one to obtain the waveforms in Figure 7. The

acceleration and velocity waveforms have large

amplitude of about 1100[cm/s2] and 150[cm/s]

and long duration of strong motion.

6. SUMMARY

A procedure for formulating Level 2 earthquake

motion is discussed and presented. Kannonzaki

site in Tokyo Bay mouth area is given as an example of target site. The Level 2 earthquake

motion evaluated here is based on the most

influential earthquake and strong motion

simulations by the stochastic Green's function

method including the effects of uncertainty of the

source parameters. The source parameters,

especially the locations of asperities, have great

effects on the simulated strong motions. An

example of Level 2 earthquake motion with large

amplitude and long duration is presented.

32
ACKNOWLEDGEMENT

In order to promote the present study, 'Technical

Committee on Methodology for Formulating

Level 2 Earthquake Motions (Chairman: Prof.

Tatsuo Ohmachi, Tokyo Institute of Technology)"

has been established at the National Institute for

Land and Infrastructure Management. The

authors would like to thank all members of the

committee for the valuable discussion.

REFERENCES:

Irikura, K. (1986): Prediction of strong

acceleration motions using empirical Green's

function, Proc. 7th Japan Earthq. Eng.

Symp., pp. 151-156.

Japan Road Association (2002): Design

specifications for highway bridges, Part V.

Seismic design (in Japanese).

Japan Society of Civil Engineers (1996): Proposal

on earthquake resistance for civil

engineering structures (in Japanese).

Kanamori, H. (ed.) (1991): Physics of earthquake

(in Japanese).

Matsuda, T. (1995): Seismic zoning map of

Japanese islands, with maximum magnitudes

derived from active fault data, Bull. Earthq.

Res. Inst., Univ. of Tokyo, Vol. 65,

pp.289-319 (in Japanese).

Nakao, Y. et al. (2002): Seismic hazard map for

all Japan, Proc. 11th Japan Earthq. Eng.

Symp. (submitting, in Japanese).

Satoh et al. (2002): Empirical models of

acceleration response spectra and

envelopes at engineering bedrock using

K-NET strong motion records, Proc. 11th

Japan Earthq. Eng. Symp. (submitting, in

Japanese).

Somerville, P. et al. (1999): Characterizing crustal

earthquake slip models for the prediction of

strong ground motion, Seism. Res. Letters,

Vol. 70, No. l,pp. 59-80.

Takahashi, T. et al. (2000): A spectral attenuation

model for Japan using strong motion data

base, Proc. 6th Int. Conf. on Seismic

Zonation, 6p. (CD-ROM)

The Research Group on Active Faults of Japan

(1991): Active faults in Japan, Rev. Ed.,

Univ.of Tokyo Press (in Japanese).

Wald, D. J. and Somerville, P. G. (1995):

Variable-slip model of the Great 1923 Kanto,

Japan earthquake: Geodetic and

body-waveform analysis, Bull. Seism. Soc.

Am., Vol. 85, No. 1, pp. 157-177.

33
Figure 1. Epicenters of major earthquakes (Mj

>=6.5) occurred around Kannonzaki site in

the past and distribution of active faults (The

Research Group on Active Faults of Japan,

1991). The locations of epicenters are

determined by Japan Meteorological Agency

(JMA).

slip |rn]

Figure 3. Map showing the rectangle asperities,

Asp#l and Asp#2, identified from the

variable-slip model of the 1923 Kanto

earthquake (Wald and Somerville, 1995) and

the location of Kannonzaki site.

Target site Engineering bedrock

Ground motions due to subfaults

Figure 2. Schematic illustration of the synthesis

of ground motions due to subfaults to produce the ground motion due to the target

source fault model.

! : .« c

. ii

its.

Figure 4. Source fault models used in the

simulation with locations of asperities

(shade) and rupture starting points (star).

34
natural period (s)

Figure 5. Acceleration response spectra of the

135 simulated waves on the engineering

bedrock at Kannonzaki site compared with

Sm for Class 1 ground of DS02.

0.1 1

natural period (s)

Figure 6. The same as Figure 5 but compared

with acceleration response spectra with 2, 5,

and 10% probability of exceedance in 100

years derived by probabilistic seismic hazard

analyses (Nakao et al., 2002). An example

of Level 2 earthquake motion formulated

based on the simulated waves is also shown.

-1200

150

o

o

-150

40 60 80

time (s)

120

20 40 60 80 100 120

time (s)

Figure 7. Acceleration and velocity waveforms of an example of Level 2 earthquake motion. One of

the simulated waves is chosen and its Fourier amplitude spectrum is modified so that its acceleration

response spectrum fits to the example of Level 2 earthquake motion shown in Figure 6.

35

Identification of Site Frequencies from Building Response Records

by

M. Celebi'

ABSTRACT

A simple procedure to identify site frequencies

using earthquake response records from roofs and

basements of buildings is presented. For this

purpose, data from five different buildings are analyzed using only spectral analyses techniques.

Additional data such as free-field records in close

proximity to the buildings and site

characterization data are also used to estimate site

frequencies and thereby to provide convincing

evidence and confirmation of the site frequencies

inferred from the building records. Furthermore,

simple code-formula is used to calculate site

frequencies and compare them with the identified

site frequencies from records. Results show that

the simple procedure is effective in identification

of site frequencies and provides relatively reliable

estimates of site frequencies when compared with

other methods. Therefore the simple procedure

for estimating site frequencies using earthquake

records can be useful in adding to the data base of

site frequencies. Such data bases can be used to

better estimate site frequencies of those sites with

similar geological structures.

KEYWORDS: site-frequency, structural

frequency, structural response, spectral analysis,

cross-spectrum, transfer function, coherence.

1 Earthquake Hazards Team, USGS (MS977),

345 Middlefield Rd., Menlo Park, Ca, 94025

1.0 INTRODUCTION

Reliable calculations and/or estimates of the

fundamental frequency of a building and its site are essential during analysis and design process.

Various code formulas based on empirical data

are generally used to estimate the fundamental

frequency of a structure. Alternatively, if

dynamic modal analysis is performed,

fundamental frequencies and mode shapes are

obtained. For existing structures, in addition to

code formulas and available analytical tools such

as modal analyses, various methods of testing

including ambient and forced vibration testing

procedures may be used to determine dynamic

characteristics. Reliable strong-shaking dynamic

characteristics are obtained, when and if the

structures are instrumented and their on-scale

responses are recorded during strong shaking

events. Spectral procedures and system

identification techniques applied to the recorded

strong shaking response data yield very accurate

assessments of the actual dynamic characteristics.

While structural frequencies can be calculated

using mathematical models or determined from

records, identification of site frequencies are not

as straightforward; hence, often, the estimation of

site frequencies are made using simple empirical

relationships with rough parametric values and

without mathematical modeling. Recent codes

provide approximate estimates of site frequencies

if geotechnical logs are available. Furthermore,

there are always some uncertainties in prediction

of site frequencies because of the assumptions

made in establishing representative site

characteristics. The frequently used simple

formula, TS=4H/VS , requires rninimal but

reasonable characterization of depth to bedrock

and representative average shear wave velocities

of layered media (International Building Code,

2000).

In this paper, the objective is to show that, when

and if structural response records are available

from both the roof and ground floors or basements of building structures, simple spectral

analyses procedures can be used to convincingly

identify not only structural frequencies but also

site frequencies.

37
The benefits of such an identification procedure

is considerable. Identified site frequencies can be

used in data bases that aim to classify site 2. characteristics against dynamic characteristics,

and, in assessing other techniques used to identifying site frequencies. Furthermore, the

procedure can be applied to sets of data available

from code-type instrumented buildings (three triaxial accelerographs placed at the roof, midheight and basement of buildings) as well as from

buildings instrumented with multiple sets of

sensors in different floors. A data base that results

from such assessments can be used for similar

sites when, otherwise, there is insufficient

information to infer site frequencies.

The scope of this paper is limited to demonstration of the procedure with five sets of

building response data, four obtained during the

1989 Loma Prieta earthquake and another

obtained during the Whittier, California

earthquake. The intent here is to identify

structural frequencies and site frequencies from

synchronized strong-shaking data recorded during

strong-shaking events from instrumented

structures. Low-amplitude test data is not used in

this study.

For all of the five cases included in this study,

detailed assessments of structural characteristics as well as assessments of site frequencies were

included in previous studies with much wider

scopes (Celebi, 1992, 1993a,b,c, 1994, 1998).

This study concentrates on primarily for

identification of site frequencies using recorded

building responses motions. It is envisioned that

data bases of site frequencies extracted from

building responses records may be developed for

future use.

2.0 THE PROCEDURE

In order to identify site frequencies, the following

steps are essential:

1. At least two pairs of horizontal

components of recorded data, one

pair from the roof and the other from

the ground floor or basement are

required. Either parallel and/or

orthogonal pairs of data from roof or basement can be used.

The structural frequencies are

identified first. The roof and/or upper

floor records are naturally the best

suited for this. The following well

known methods are used to identify

structural frequencies:

a. spectral analyses (amplitude

spectra and spectral ratios, cross- spectra and coherence and phase

relationships) and

b. system identification methods. In

the event that system

identification procedures are

used, the roof and/or upper floor

data constitute the output and

basement and/or ground floor

records are adopted as the input

motion.

Once structural frequencies are

confidently identified, then the site

frequency distinguished:

a. if one of the non-structural

dominant peaks of cross- spectrum of ground-floor (or

basement) motions is different

than the structural frequency,

than that frequency is likely the

site frequency,

b. if the spectral ratio cancels out a dominant frequency that clearly

appears in amplitude and/or

cross-spectrum, then that

frequency is not a structural

frequency but it most likely is the site frequency, and/or,

c. cross-spectra or normalized

cross-spectra [Sxy/max(Sxy)]

calculated from pairs of roof and

basement (or ground floor) data

exhibit site frequencies.

Availability of free-field records

from a free-field station that is in the

proximity of the building adds

further confidence in confirming the

identified site frequency. The

amplitude spectra of the components

of and/or cross-spectrum of

orthogonal horizontal components of

38
free-field motions usually reveals the

site frequency.

5. If, in addition, site characterization

data (depth to bedrock, geological

borehole data, shear-wave velocities

of different layers of soil below the

foundation) is available, transfer

functions can be calculated to add

further confidence.ions can be

calculated to add further confidence.

2.1 Case 1: Pacific Park Plaza [PPPJ,

Emeryville, California:

Detailed analyses of 1989 Loma Prieta

earthquake (Ms=7.1) response recordings of the

30-story, reinforced concrete framed Pacific Park

Plaza [PPP] in Emeryville, California, have been

presented by Celebi and Safak (1992), Anderson,

Mirando and Bertero (1991) and Celebi (1998).

Recently compiled borehole and site

characterization information is also available

(Gibbs, Fumal and Powers, 1994). Figure 1 shows plan view, instrumentation scheme of the

building and the location of the free-field stations

associated with this building instrumentation. The

available free-field strong motion recording is pertinent to the convincing identification of the

site frequency. Figures 2a and b depict building

accelerations recorded at the core of the top

instrumented level, at the core of the ground floor

and the associated south free-field of the threewinged building. Corresponding amplitude

spectra are provided in Figures 2c and d. The first three modal structural frequencies (periods)

clearly identified from the recordings are 0.38,

0.95 and 1.95 Hz (2.63, 1.05, 0.34 s). The peak at 0.7 Hz that appears in the amplitude spectra of

the roof also appear as the dominant peak in the

amplitude spectra of the ground floor and the

south free-field (SFF). However, this peak at 0.7

Hz disappears in the spectral ratios calculated

from the amplitude spectra of the roof and ground

floor as depicted in Figures 2e and f. This

indicates that 0.7 is the site frequency as, although it appears in the roof spectra, it cancels

out when ratios are calculated.

In Figure 3, cross-spectra, calculated from pairs

of orthogonal components of acceleration

recorded at the (a) roof , (b) gound floor and (c)

free-field are presented. The roof cross-spectrum

clearly identifies the aforementioned frequencies

of the first three modes. These modes are coupled

torsional-translational modes (Celebi, 1998). The

peak at 0.7 Hz that appears in the cross-spectrum

of the roof appears as the dominant peak in the

cross-spectra of the ground floor and the south

free-field (SFF). Next, when the normalized

cross-spectra are calculated for the ground floor

and free-field, the site frequency at 0.7 Hz is distinguishable from the structural frequencies in

the normalized cross-spectrum of the roof (Figure

2d). This is further confirmed by the lowest

frequency peak at 0.7 Hz of the transfer function

(Figure 4) calculated by using Haskell's shear

wave-propagation method (Haskell, 1953, 1960)

using site characterization data related to

variation of shear wave velocities with depth

(Gibbs, Fumal and Powers, 1994). The depth to

bedrock has been adopted from a map by Hensolt

(1993) as 150 m. (-500 ft).

2.2 Case 2: Two-story Office Building

[OAK], Oakland, California:

McClure (1991) provides detailed particulars of

the two-story office building in Oakland,

California. The instrumentation scheme of this

building as well as accelerations recorded during

the 1989 Loma Prieta earthquake from the roof,

second and ground floors are provided in Figure

5. Ambient tests of the building performed in the

1965 yielded first mode frequency (period) as 2.13 Hz (0.47 sec) and 2.08 Hz (0.48 sec) for the

NS and EW, respectively, and forced vibration

tests, also performed in 1965, yielded 2.35 Hz

(0.426 sec) (Bouwkamp and Blohm, 1966).

These and other assessments of modal

frequencies of the building are summarized in

Table 1.

Figure 6 shows amplitude spectra of recorded

accelerations in both the NS and EW directions

and rotational accelerations (difference between

parallel records) at the three structural levels

(roof, second floor and ground floor). Figure 7

shows time-histories, amplitude spectra and

spectral ratios for pairs of recorded accelerations

at the roof and ground floor. From the spectra,

39
three distinctive frequencies (0.82-0.85, 1.65 and

1 .95 HZ) are identified. The frequencies 1 .65 Hz

and 1 .95 Hz are structural frequencies determined

by the fact that they have a very high ratio

amplitude as seen in the in the spectral ratio plots

(Figure 6g-6i) calculated from the pairs of

amplitude spectra of the roof and ground floor

motions (Figure 6d-6f). These two frequencies

are very close to one another. Therefore, given

the structural irregularity created mainly by the

north and east end walls, the structure responds in a closely-coupled translational-torsional mode

with frequency between 1 .65-2 Hz. The 0.82-0.85

Hz (NS) and 0.65-0.85 Hz (EW) frequencies that

appear in the amplitude spectra do not appear in

the spectral ratios because they cancel out. Therefore, it is safe to declare that the site

frequency is between 0.65-0.85 Hz.

Figure 8 depicts coherence, phase angle and

cross-spectrum of the pairs of parallel motions at the roof [CH2 and CH3] and ground floor [CH6

and CH7]. Because the 2-story building is very

rigid, both the structural frequency and the site

frequency appear in the cross-spectrum plots of

the roof and ground floor motions, although the

amplitude of the site frequency at the ground

floor is much larger than that of the structural

frequency.

There is no free-field station associated with the

building; however, recently documented site

characterization data in proximity to the building

(Gibbs, Fumal and Powers, 1993) allows

determination of site transfer function (Figure 9).

Depth to bedrock (two cases) have been

estimated from Hensolt map (1993). There is a good match between the lowest frequency (0.6-

0.7 Hz) in Figure 9 and the site frequency (0.65-

0.85) extracted from the building records (Figures

6-7).

2.3 Case 3: Santa Clara County Office

Building [SCCOB], San Jose, Ca.

The building for Case 3, Santa Clara County

Office Building (SCCOB) in San Jose, California

is perhaps the most complex response cases caused by three close frequencies (0.38, 0.45, and

0.57 Hz) (Celebi, 1998). Figure 10 depicts the

instrumentation scheme and the relative location

of the building and the epicentral locations of the

three earthquakes that were recorded on and

before October 17, 1989 Loma Prieta, Ca.

earthquake.

Figure 1 1 shows the very unique responses of the

roof of the building to the three different

earthquakes. These are typical exhibitions of

coupled torsional and translational responses with

significant beating effect caused by closely

spaced translational frequency (0.45 Hz) with the

torsional frequency (0.57 Hz) and low critical

damping of approximately 2 % of the structural

system (Celebi, 1994, 1998, Boroschek and

Mahin, 1991). Due to this type of behavior, the

building was retrofitted by adding viscous elastic

dampers (Crosby, Kelly and Singh, 1994).

Although strong shaking data has not been

recorded since the retrofit in 1994, it is expected

that in the future the response of this building will

not resemble those in Figure 11 and due to

expected shift in building frequency and

increased damping, the beating effect will

disappear.

Figures 12a-c shows pairs of parallel translational

accelerations at the roof and their differences

representing torsional accelerations (NS: CH6,

CH7, and CH6-CH7) and (EW: CH4, CH5 and

CH4-CH5) and translational accelerations (NS:

CH22, EW: CH20 and CH21) at base of the

SCCOB. Corresponding amplitude spectra of

these motions are provided in Figure 12d-f. The

fundamental frequency (period) of the building at 0.45 Hz (2.22 sec) belongs to the translational

mode and the frequency (period) at 0.57 Hz

(1.75 s) belongs to the torsional mode; hence, the

closely coupled translational-torsional response

of the building that causes the beating effect.

Details of these effects are provided by Celebi

(1994, 1998) and Boroschek and Mahin (1991).

The frequency at 0.38 Hz (2.63 sec) belongs to

the site.

The attributions to structural and site frequencies

are confirmed by the spectral ratios of roof

motions with respect to base motions (Figures 13 a and b). The site frequency (0.38 Hz) cancels out

in the spectral ratios of roof/base motions. This

40
frequency (0.38 Hz) also appears in the cross- spectra of orthogonal (CH21 and CH22) and

parallel (CH20 and CH21) pairs of motions at the

base; hence, indicating that it is site related and

not structural related.

Limited geological logs (Earth Sciences, 1971)

available allows approximate calculation of site

transfer function using estimated shear wavevelocities and depth to bedrock estimated

anywhere between 150-270 m. Figure 14 shows

that the site, in the Santa Clara basin, is capable

of generating motions with periods between 2-3

seconds, depending on the assumed depth to

bedrock. The long-period site characterization is also confirmed by a study of the basin effect in

the Santa Clara (CA) by Frankel and Vidale

(1992) who concluded that 2-5 second longperiod motions can be generated in this particular

basin.

2.4 Case 4: Embarcadero Building [EMBJ,

San Francisco, Ca.

Figure 15 shows a three-dimensional view and

the instrumentation scheme of the Embarcadero

Building (EMB) in San Francisco, Ca.

Figures 16a-b show the acceleration responses

recorded at the roof and basement of EMB in the

NS and EW directions, respectively. The

normalized amplitude spectra of these motions

are depicted in Figures 16c-d. The reason these

spectra are normalized is to show the significant

peaks of both the roof and basement in the same

plot. Otherwise, since the building is tall, the

basement spectra would not clearly be seen if

they are plotted on an equal scale. The

fundamental frequencies (periods) of the building

are identified as 0.19 Hz (5.26s) in the NS and

0.16 Hz (6.25s) in the EW directions respectively.

Detailed analyses of recorded data from this

building is presented by Astaneh, Bonovitz and

Chen (1991), and Celebi (1993a). Figures 16e-f

show the ratios of amplitude spectra of pairs of

roof and basement motions. The site frequency

(period) at 0.7-0.8 Hz (1.25-1.43 sec) clearly seen

in the normalized amplitude spectra of basement

motions cancels out in the spectral ratio plots. In

the spectral ratio plots (Figure 17 e-f), the second

(NS: 0.57Hz [1.75 s], EW: 0.46 Hz [1.02 s]) and

third (NS; 0.98 Hz and EW: 0.77 Hz) modal

frequencies are clearly identifiable. These

periods, in general, follow the T, T/3, T/5 rule-ofthumb.

Figure 17 shows coherence, phase angle and

cross-spectrum plots of the pairs of motions at the

roof (NS: CH 1 7 and CH1 8) and at the base (NS:

CH4 and CH6). Cross-spectrum of the pair of

roof motions, shown in Figure 1 7e clearly

indicates the third modal frequency (0.98 Hz).

The same for basement motions has a much wider

frequency band that includes the 0.98 Hz

frequency. This implies that the site frequency

(-0.7-0.9 Hz) that appears in the amplitude

spectrum of the basement motions is very close

to the third modal frequency in the NS direction

and possibly causes resonance.

The site frequency is 0.7-0.8 Hz (1.25-1.43 sec)

and is identifiable in the cross-spectra. The site

transfer function, presented in Figure 19,

confirms this identification. Site characterization

data has been adopted from Gibbs and others

(1994).

2.5 Case 5: Norwalk Buildings,

Norwalk, Ca. [NOR]

Figure 1 9 shows the extensive instrumentation for

the two buildings and the site at 12400 block of

Imperial Highway, Norwalk, Ca. In this study,

only Building B (hereby called NOR) and the

south free-field (SFF) is considered. Responses of

both buildings and three of the four free-field

stations were recorded during the 1987 Whittier,

Ca. earthquake (Ms=5.8). Detailed studies of

these records are reported elsewhere (Celebi,

1993b and c).

Figure 20 shows (a,b) NS and EW accelerations

at the roof and the basement ofNOR and its south

free-field. Figures 20c and d show the

corresponding amplitude spectra calculated from

these accelerations. Figures 20e anf f show the

corresponding spectral ratios calculated from the

amplitude spectra. The fundamental structural

frequencies of Building B are identified as 0.76

Hz (NS) and 0.83 Hz (EW). The very small peak

41
at 0.3 Hz observed in the amplitude spectra as well as in the basement cross spectra is the site

frequency. It is noted that this frequency cancels

out in the spectral ratio plots (Figures 20e and f)- Another possible larger mode site frequencies is

detectable at approximately 1.7 Hz. This

frequency also cancels out in the spectral ratio

plots.

The depth to bedrock in the vicinity of the

buildings is not well described; however, logs

from existing nearby oil wells indicate that below

500 m depth, well-cemented sandstone and

marine rock are prevalent. The site transfer

function using estimated shear wave velocities is shown in Figure 21 and exhibits that the site is capable of generating motions with low

frequencies (e.g. 0.3 Hz).

3.0 ESTIMATES OF SITE FREQUENCY

USING CODE FORMULA:

In order to assess the reliability of the extracted

site frequencies and to facilitate comparisons, the

approximate site formula (TS=4H/VS) is used to

calculate approximate periods of the sites using

the site characterization data displayed in the site

transfer function plots (Figures 4, 9, 14, 18 and

22). These calculations are summarized in Table

2 and compared with the site frequencies assessed

from earthquake records and site transfer

functions. It is noted herein that the above simple

formula uses an average shear wave velocity,

Vs(ave)=H/( I (h/Vsi) and that while the Uniform

Building Code (1997) does not state a limitation

on the total depth for which this formula can be

used, the new International Building Code (2000)

limits the use of this relationship up to a depth of ~ 30m (100ft).

4. DISCUSSION AND CONCLUSIONS

A procedure that is used to identify site frequency

from building responses recorded during

earthquakes has been presented. The procedure

distinguishes site frequency among several

frequencies that appear in amplitude spectra and

cross-spectra of horizontal records from roof and

basements of buildings. These results are further

substantiated by analyses of additional data from

associated free-field records of each building and

site transfer function calculated by using site

characterization data. Furthermore, these results are compared with simple code formula

computation even though the formula, T=4H/Vs,

may be limited to layered media with depths to 30

meters (100ft) (International Building Code,

2000). All results presented for five cases are summarized in Table 2 and comparatively plotted

in Figure 23. As noted both in Table 2 and Figure

23, the site frequencies identified from building

records are higher than those estimated by

transfer functions and the simple code formula.

The procedure can be used to process numerous

sets of accumulated data from instrumented

structures and promises to be an effective and

simple technique to identify site frequencies from

actual building responses recorded during

earthquakes.

The only real difficulty in applying this procedure

could arise when and if the site and structural

frequencies are identical or if they are too close to

one another; in which case, further examination

of the data by other procedures can be applied or

estimates of free-field data or site data can be

used to clarify the situation.

5.0 REFERENCES

1 . Anderson, J. C, Miranda, E., and Bertero, V.

V., 1991, Evaluation of the seismic

performance of a thirty-story RC building,

University of California, Berkeley,

California, Earthquake Engineering Research

Center Report 91/16, December 1991.

2. Astaneh, A., Bonowitz, D., and Chen,

C.,1991, Evaluating design provisions and

actual performance of a modern high-rise

steel structure. SMIP91 Seminar on

Seismological and Engrg. Implications of

Recent Strong-Motion Data, California

Department of Conservation, Division of

Mines and Geology, 5-1-5-10.

42
3. Boroschek, R. L., and Mahin, S. A., 1991,

Investigation of the seismic response of a

lightly-damped torsionally-coupled building,

University of California, Berkeley,

California, Earthquake Engineering Research

Center Report 91/18, December 1991.

4. Bouwkamp, J. G., and Blohm, J. K., 1966,

Dynamic response of a two story steel frame

structure, Bull. Seismological Society of

America, 56, 6.

5. Celebi, M., and Safak, E., 1992, Seismic

response of Pacific Park Plaza. I: Data and

preliminary analysis, J. Struct. Engrg.,

ASCE, 1 18(6), 1547-1565, June 1992.

6. Celebi, M., 1993a, Seismic response of

eccentrically braced tall building, J. Struct.

Engrg., ASCE, 119 (4), 1188-1205, April

1993.

7. Celebi, M., 1993b, Seismic responses of two

adjacent buildings. I. Data and analyses, J.

Struct. Engrg, ASCE, 119(8), 2461-2476,

August 1993.

8. Celebi, M., 1993c, Seismic responses of two

adjacent buildings. JX Interaction, J. Struct.

Engrg, ASCE, 119 (8), 2477-2492, August

1993.

9. Celebi, M., 1994, Response study of a

flexible building using three earthquake

records, ASCE, Structures Congress, Atlanta,

Georgia, April 1994.

10. Celebi, M., 1998, Performance of Building

structures - A Summary, in The Loma Prieta,

California, Earthquake of October 17, 1989 -

Building Structures (M. Celebi, editor),

USGS Prof. Paper 1552-C,pp. c5-c76,

January 1998.

11. Crosby, P., Kelly, J., and Singh, J. P., 1994,

Utilizing visco-elastic dampers in the seismic

retrofit of a thirteen story steel-framed

building, in Structures Congress XJJ,

Atlanta, Ga., v. 2: American Society of Civil

Engineers, p. 1286-1291.

12. Earth Sciences Associates, 1971, Soil and

foundation investigation for the proposed

Santa Clara County Civic Center, Technical

Report, Earth Sciences Associates, Palo Alto,

Ca., November 1971.

13. Fumal. T. E., 1991, A compilation of the

geology and measured and estimated shearwave velocity profiles at strong-motion

stations that recorded the Loma Prieta,

California, earthquake, U.S.G.S. Open-File

Report 91-311, Menlo Park, Ca.

14. Frankel, A., and Vidale, 1992, A threedimensional simulation of seismic waves in

the Santa Clara Valley, California from a Loma Prieta aftershock, Bull Seism.. Soc. of

America, 82, 5, pp. 2045-2074.

15. Tamura, Y., Matsui,M. Pagnin, L-C, and

Yoshida, A., 2001, Measurement of Windinduced Response of Buildings using RTKGPS, Proceedings of the 5th Asia-Pacific

Conference on Wind Engineering, Kyoto,

October 21-24, 2001 (to be published).

16. Gibbs. J. F., Fumal, T. E., and Powers, T. J., 1993, Seismic velocities and geological logs

from borehole measurements at eight strongmotion stations that recorded the 1989 Loma

Prieta, California, earthquake, U.S.G.S.

Open-File Report 93-376, Menlo Park, Ca.

17. Gibbs, J. F., Fumal, T. E., Borcherdt, D. M.,

Warrick, R., Liu, H-S, and Westerlund, R.,

1994, Seismic velocities and geologic logs

from borehole measurements at three

downhole arrays in San Francisco, California,

U.S.G.S. Open-File Report 94-706, Menlo

Park, Ca.

18. Gibbs, J. F., Fumal, T. E., Boore, D. M., and

Joyner, W. B., 1992, Seismic velocities and

geologic logs from borehole measurements at seven strong-motion stations that recorded

the Loma Prieta earthquake, U.S.G.S. OpenFile Report 92-287, Menlo Park, Ca.

19. Gibbs, J. F., Fumal, T. E., and Powers, T. J., 1994, Seismic velocities and geologic logs

from borehole measurements at seven strongmotion stations that recorded the 1989 Loma

Prieta, California, earthquake, U.S.

Geological Survey, Open-File Report 94-222,

Menlo Park, Ca., 104 pages.

20. Hensolt, W. H., 1993, Central San Francisco

Bay Region bedrock contour map, in

preparation, personal communication.

43
21. Haskell, N. A., 1953, The dispersion of

surface waves on multi-layered media, Bull.

Seismological Soc. Am., 43(1), 17-34.

22. Haskell, N. A., 1960, Crustal reflection of

plane SH waves, J. Geophysical Res.,

65(12), 4147-4150.

23. McClure, F. E., 1991, Analysis of a two-story

Oakland office building during the Loma

Prieta earthquake, SMIP91:Seminar on

Seismological and Engineering Implications

of Recent Strong-Motion Z)a/a,California

Department of Conservation, Division Mines

and Geology, 13-1-13-13-11, May 1991.

Table 1 . Modal Frequencies (Periods) of the 2-Story Oakland Building

Assessment Method Modal Frequencies [Hz]

(Periods [seconds])

Comments

NS EW

1 965 Ambient Test

(McClure, 1991)

2.13 (1

st )

(0.47)

2.08

(0.48)

Walls in, no plaster

1965 Forced Vibration Test

(Bouwkamp & Blohm,1966)

(1

st) 2.35 (0.426),

(2

nd

) 7.70 (0.130)

Walls in with plaster

1988 UBC (McClure, 1991) 2.29 (0.437) Code Formula

McClure Computer Model

(1991)

(1

st) 1.59 (0.63),

(2

nd

) 5.0 (0.20)

w/o plaster

(1

st) 2.16 (0.463),

(2

nd

) 7.69 (0.130)

With plaster

Spectral analyses (this study) 1.95(0.51) Translational,

1.65 (0.61) Torsional

Loma Prieta (1989)

data

44
Table 2. Assessment of site period (frequency) using code formula [Vs(ave)=H/( I (hj/VSj)],

earthquake records and transfer functions.

Method Parameter PPP OAK 2ST SCCOB EMB NOR

A B A B

H=Shj [m] 150 152 169 270 500 64 500

Formula I (h/VsO 0.46 0.41 0.46 0.66 1.05 0.33 1.01

Vs(Ave)

[m/s]

329.5 366.5 369.6 407.8 478.3 195.8 493.7

Site

T=4H/Vs [si

1.82 1.65 1.83 2.65 4.18 1.31 4.05

Site M/t

[Hz]

0.55 0.60 0.55 0.38 0.24 0.76 0.25

Records Structural

T(s)

2.63 0.51 (trans.),

0.61 (tors.)

2.22(trans.),

1.75(tors.)

5.26(NS)

6.25(EW)

1.32(NS)

1.20(EW)

Structural

f(Hz)

0.38 1.95(trans.),

1.65(tors.)

0.45(trans.),

0.57(tors.)

0.19(NS)

0.16(EW)

0.76(NS)

0.83(EW)

SiteT(s) 1.33 1.18-1.25 2.63 1.11-1.42 3.3

Site f (Hz) 0.7 0.8-0.85 0.38 0.7-0.9 0.3

Transfer

Function

Site T (s) 1.18-

1.54

1.43-1.54 2.0-4.0 1.25-1.42 3.3

Sitef(Hz) 0.65-

0.85

0.65-0.70 0.25-0.5 0.7-0.8 0.3

Figure 1. General Layout and Instrumentation Scheme of Pacific Park Plaza, Emeryville, Ca.

45
LOMAPHETAFFFpSO) UMMPnETAPPP(260}

Figure 2. (a,b) Recorded orthogonal accelerations at the roof, ground floor and south free-field of

Pacifica park Plaza, (c,d) corresponding amplitude spectra and (e,f) ratios of amplitude spectra.

15000

10000

PPR GROUND FL

1 2 3 4 5

X10

4 PPP: SOUTH FF

3 0.7

A350&A260

1 2 3 4 5

FREO (HZ)

0.7 ;

A3SO&A260

H

LLL

0 1 2 3 4 5

PPR ROOF, QR.FL.SFF

H

OF

FL

.7jU A35O&A260

1 2 3 4 5

FREQ(HZ)

Figure 3. Cross-spectra of motions at the (a) roof, (b) ground floor, and (c) free-field of Pacific Park

Plaza, Emeryville, California, and (d) normalized cross-spectra.

SITE TRANSFER FUNCTION: PACIFIC PARK PLAZA

84

2

I

! ! A bEFTHflil)

\ f\\ 1\ <«.s :

: / V / : \ 25-S.2 i 0279

0.143

0.067

1/ : \ 8.1-S3S 0241

0/402

0280

/ ; — i 51.M0up \

0-5 1 1.5 2 25 3 3,5

FRBQ(HZ)

Figure 4. Site transfer function for described characterization at Pacific Park Plaza.

46
0 8 10 18 20 0303840

TIE(8] Figure 5. (Left) Instrumentation scheme of the torsionally eccentric 2-story building in Oakland, Ca.

and (Right) Accelerations recorded at the roof, second and ground floors during the 1989 Loma

Prieta, Ca. earthquake.

no

4 M x1tf ROTATIONAL

5

2 4

FREQQiZ}

1.95 !

CH2-CH3

2 4

I IUJQ (HZ)

Figure 6. Amplitude spectra oftranslational and torsional accelerations recorded at Oakland (2-

Story Building).

ItCHM EWCH2 4CH8 EWiGMftCHF

.b4 * : if]

k CHS

Figure 7. [a-c] Time-histories ofroof and ground floor acceleration pairs, [d-f] corresponding

amplitude spectra and [g-i] corresponding rooffground floor spectral ratios.

47
0MCEW(CKZiOe) OAK£W(CHB4CH7)

4 S 4 9 M ;

100 —,

«1lT

1 2

Figure 8. [a,b] Coherence, [c,d] phase and [e,f] cross-spectra plots of pairs of parallel motions at the

roof (CH2 and CH3) and ground floor [CH6 and CH7].

SITE TR/WflFEfl FUNCT!Of«MiaLWC^8n3RY BLHLCHNQ

Figure 9. Computed site transfer function for described site characterization (see figure) ofthe site

ofthe 2-story building, Oakland, California.

V *

... ^ t.

r— 1 1 . > n.i*bi I * I'lH^O -n.- • * - Ikf tt -S »M .2

I

J' - ^

^

'l

\\1

i

i Li**

-

: s v.. > i.

v. 'V: Ui^HI^^

' •• ,»w

- K**

Figure 10. Location map of Santa Clara County Office Building in San Jose, Ca. and the epicenters

of the three earthquakes that were recorded by the instrumentation array within the building.

48
RESPONSE AT ROOF OF 900QB: CH8(N9): BEFORE RETROFIT

U3MA PRETA EO4IMS)

20 40 6680

*-

00 T?fiS

TIME (S) RESPONSE AT ROOF OF SCCOB: CH5(EW): BEFORE RETROFIT

Tao

MORGAN HILL EQ.(1884)

05 -^m^\f^^^ L.

. Li j i i . MT.LEWB;EQ.(18S0) j

ll" : : " UOMA L PRIETA F EOJ190) EQ^tOtS)

20 40 & rtH" 120

TIME (8) Figure 11. Time histories of accelerations recorded at the roof of the Santa Clara County Office

Building during the 1989 Loma Prieta, 1996 Morgan Hill and 1994 Mt. Lewis earthquakes. The

building was retrofitted in 1994. The responses exhibit beating effect caused by closely couple

translational and torsional modes and low damping.

LPG8000B-ROOF LPErSCCOB-flOOF LPE8O0OB-BA8E

22

3 ^fa.&sSS.

FRBQKHZ) FRECKHZ) FREQfHQ

Figure 12. [a-c]Translational and torsional accelerations at the roof and translational accelerations at

the basement and [d-f] their amplitude spectra.

49
INS] : R00F/9A8E £EW]:R00fiBA8e

25

20 I"*10

8

0

8000

•OMCH8CH22

57

30

28

<X6 1 1.6 2

FREQ(HZ)

BASE.-CH21 V8.CH22

.48 ! lira

J0l6 1 i£ 2

FHEQ(HZ)

BASE.-CH20VS.CHei

1000

0,

0,38. 0000

2000

040 j M

I: KtCWW

OLB 1 1.8 2

FREQfrB)

08 1 14 2

FREOfHZ)

Figure 13. (a and b) Spectral ratios of roofTbasement motions indicate structural frequencies

(translational [0.45 Hz], and torsional [0.57 HZ]) and cross-spectrum ofbasement motions indicate

site frequency (0.38 HZ).

au

I

4 i"

I

' u

fu

1

affETHWWUIRIWeTI0W!tCC0ete8T»IICTEPP<>I WML I UM 1 1 1 1 1 1 1 1 1

0 Oft

(HZ)

Figure 14. Site transfer functions for 2 postulated depths to bedrock (SCCOB) site.

Figure 15. Three-dimensional schematic of Embarcadero Building [EMB] and its instrumentation

scheme.

50
1

03

0

0.5

30

I"

0,

44THFL.

CHI7

BASByBJT CH4

60 100

[el CH17Twfld]

.19

*---•.

! j 1^

OS 1 13

.19 CH17/CH4 i M

.J7.....I i

J

1 A*

03 1 13 2

(HZ)

20

10

0

Lie chiwchs : n

OS 16 2

(HZ)

Figure 16. [a-c] Time-histories of roof and ground floor acceleration pairs of EMB in theNS and

EW directions, [d-f] corresponding amplitude spectra and [g-i] corresponding roof/ground floor

spectral ratios.

EMR0t8)CH17lCH16

1

1 03

0,

M

0.6 16

1

OS

•i

1000

-100

&R(N8)CH4*CH6

m rv

BASEMENT

OS 13 2

fcfl

0 , 0J

x10»

13 03 13

0 03 1 13 2

FREO04Z)

2000

10000

n A 030-030

f

i

ni £Z. i \-'"TN

-*\-

0 03 1 13 2

FREOflHZ)

Figure 17. [a,b] Coherence, [c,d] phase angle and [e,f] cross-spectra plots ofpairs ofparallel

motions at the roof (CH 1 7 and CH1 8) and basement [CH4 and CH5]

.

EMBARGADERO BUILDING: STTE SITE TMN8FER FUNCTION

L

i

I*

i§L 8b!.. j

i f

\: 20.0-33.8 0.166 :

/ \ 0207

/ j\ 8i5.i«.a/^\a>Bi j

i i i i i i 0 03 1 1.5

Figure 18. Site transfer function for EMB site

2 25 3

FREQ(HZ)

33 45

51
•Unit

PUILOING » ./

.WW

B. ... —=J—

Figure 19. Instrumentation of the two buildings and the site at 12400 block of Imperil Highway,

Norwalk, Ca. In this sudy, Building B (NOR) and the south free-field (SFF) is considered.

WHTmERMOfWALX(NB} WHITTIBttJOHWAUCCEW)

3

2

10

I

15

10

5

0

[ej

oeiycHgr

i (

.

k 1

•

2 3

FHEO(H2)

2 3

FREOCHZJ

Figure 20. [a-c] NS and EW time-histories ofroof, basement and SFF accelerations ofNOR

building, [c-d] corresponding amplitude spectra and [g-i] corresponding roofground and roof7freefield spectral ratios.

52
[a] WHITTIERMORVKMX [b] WHITTI ER:NORWALK

FHEQ(HZ) FFEQ(HZ)

Figure 21. [a] Normalized cross-spectra ofpairs of parallel NS accelerations [CH21 and CHI 3] at the roof and EW accelerations [CHI and CH3] at the roof and second floor, and cross-spectrum of

[b] parallel NS accelerations [CH5 and CH9] at the base, [c] orthogonal accelerations [CH 21 and

CHI] at the roof and [d] orthogonal accelerations at the base [CH8 and CH9].

Figure 22. Site transfer function for NOR site.

0.8

Q0.7

~0.6

S 05 E

u-0.4

0.3

0.2

FREQUENCY VS. DEPTH

: o

-X- -

- x x

:+ FR^ REOORPS i

j o from transfer Function

i X FROM FORMULAJ

60 100 ISO 200 250 300 350

DEPTH (M)

Figure 23. Variation of site frequency with depth using 3 methods.

400 460 600

53

Evaluation of Earth Pressure Acting on Embedded Pile-cap Based on

Liquefaction Test Using Large-scale Shaking Table

by

Shuji Tamura1

, Chikahiro Minowa2

, Kohji Tokimatsu3, Shunji Fujii

4

, Kaeko Yahata5 and Tomio Tsuchiya6

ABSTRACT

Shaking table tests are conducted using a largescale laminar shear box to investigate the effects

of non-liquefied crust overlying liquefied soils

on an embedded footing. It is shown that (1) The

total earth pressure before liquefaction is induced mainly by the inertial force of the building. The shear force at the pile heads corre- sponds to the difference between the total earth

pressure and the inertial force; (2) The total earth

pressure after liquefaction is induced mainly by

the soil deformation. The shear force at the pile

heads corresponds to the sum of the total earth

pressure and the inertial force of the building;

(3) The relation between the relative displacement and the total earth pressure is linear before

liquefaction. It becomes nonlinear with the development of pore water pressure and the total

earth pressure decreases with cyclic loading after liquefaction; (4) The peak value of the total

earth pressure for the super-structure with a low

natural frequency is larger than that with a high

natural frequency. This is probably because the

inertial force of the super-structure with a low

natural frequency may interrupt the response of

the footing that tends to move with the ground.

KEY WORDS: Seismic Earth Pressure

Liquefaction

Pile

Shaking Table Test

1. INTRODUCTION

During the 1995 Hyogoken-Nambu earthquake,

extensive soil liquefaction occurred on the re- claimed land areas of Kobe. Many damaged

piles in such liquefied soil have been reported. It indicates that the effects of liquefaction on piles

should be taken into account in foundation design.

Soil-pile-structure interaction during liquefaction has been studied by field investigations

(Oh-Oka et al, 1998), numerical analyses (Fujii

et al., 1998, Tokimatsu et al., 1998), centrifuge

tests (Sato et al., 1995) and large-scale shaking

table tests (Tamura et al., 2000). Most research

efforts of these studies have focused on the

evaluation of soil-pile interaction including p-y

relations during soil liquefaction. Thus, knowledge of the effects of non-liquefied crust overlying liquefied soil on an embedded footing re- mains limited. The evaluation of the kinematic

force acting on an embedded footing is an important consideration in the seismic design

method using p-y curve for pile foundations.

This paper investigated earth pressure acting on

an embedded footing of a building during soil

liquefaction, using a large-scale laminar shear

box. The objects of this paper are; 1) to study

major factor affecting earth pressure; 2) to study

the relation between the relative displacement

1 Associate Professor, Shinshu University, 4-17-1 Wasasato, Nagano-shi, Japan, 380-8553

2 Scientific Research Advisor, National Research Institute for Earth Science and Disaster Prevention, 3-1

Ten-nodai, Tsukuba-shi, Japan, 305-0006

3 Professor, Tokyo Institute of Technology, 2-12-1 O-okayama, Meguro-ku, Tokyo, Japan, 152-8552

4 General Manager, Taisei Corporation, 344-1 Nase-cho, Totsuka-ku, Yokohama-shi, Japan, 245-0051

5 Chief Research Engineer, Kajima Technical Research Institute, 2-19-1, Tobitakyu, Chofu-shi, Japan,

182-0036

6 Chief Researcher, Takenaka Research & Development Institute, 1-5-1 Inzai-shi, Chiba, Japan, 270-

1395

55
and earth pressure; 3) to study phase difference

between the inertial force of the building and

earth pressure, and 4) to study the effects of a

natural frequency of the super-structure on earth

pressure.

2. MODEL PREPARATION

Shaking table tests were performed at NIED

(National Research Institute for Earth Science

and Disaster Prevention) in Tsukuba, Japan. A

large-scale laminar shear box 6 m high, 3.5 m

wide and 12 m long (shaking direction) was

mounted on the shaking table. A sketch of a

large-scale laminar shear box including a soilpile-structure system for Model B-S is presented

in Fig. 1 . Four steel piles with a diameter of 16.5

cm, a length of 5.3 m, a thickness of 3.7 mm and

EI=1259 kNm were installed in a three-layer

soil profile with a thickness of 6m. This profile

included a 4.5 m layer of Kasumigaura sand

(emax=0.96, ^=0.57, D50=0.31 mm, Fc =5.4%)

with a shear wave velocity of 90 m/s, placed on

top of a 1 .5 m layer of gravel with Vs = 230 m/s.

The water level was about GL-0.5 m, that is, 0.5

m dry sand overlying 4 m saturated sand. The

predominant frequency of the ground was about

5 Hz. The super-structure model with a mass of

14200 kg was supported by 4 vibration isolation

rubbers, 4 laminated rubber bearings and 2 vis- cous dampers fixed on a steel footing (2.5m

long, 1.8m wide and 1.6m high). The footing

with a mass of 2100 kg was embedded 0.5 m in

the dry sand. A mass of the super-structure is about 6 times that of the footing. The superstructure had a natural frequency of about 5.4

Hz, which was larger than the predominant frequency of the ground. The pile heads were rigidly linked to the footing, while their tips were

connected to the laminar shear box by hinges.

The test set up of Model B-L is similar to that of

Model B-S, except that the super-structure was

supported by 4 laminar rubber bearings and 2

viscous dampers only. The natural frequency of

this super-structure was about 1.3 Hz, which was

lower than the predominant frequency of the

ground.

Both models, B-S and B-L, were excited by

PJNKAI92, which is a synthesized ground motion for the Tokyo Bay area expecting the re- occurrence of the 1923 Kanto earthquake. The

amplitude of the motion was scaled as 240 cm/

s . Acceleration, displacement, excess pore water pressure and strain of the piles were recorded

during the tests.

3. TEST RESULTS AND DISCUSSION

3.1 Dynamic Response of Soil-Pile-Structure

System

Figure 2 shows the acceleration time histories of

the super-structure, the ground surface and the

shaking table for Models B-S and B-L. Figure 3

shows the time histories of bending moment at the pile heads for both models. The bending

moment begins to increase rapidly at 8 seconds

in both models. The acceleration of the superstructure, in contrast, begins to increase at 2

seconds as shown in Fig. 2. Figure 4 shows the

time histories and the vertical distribution of the excess pore water pressure. The excess pore

water pressure begins to increase rapidly at 8

seconds in both models. This corresponds to the

time when the bending moment at the pile heads

begins to increase rapidly. This suggests that the

bending moment is apparently influenced by

soil liquefaction.

Soil liquefaction progressed from the upper

layer to the bottom. The lower part of the saturated sand layer almost liquefied at about 20

seconds. The ground acceleration and the development of the excess pore water for Model BS are roughly the same as that of Model B-L.

This indicates that the soil response for Model

B-S is similar to that of Model B-L. The acceleration of the super-structure, however, contains

higher frequency components in Model B-S than

in Model B-L until 8 seconds. The predominant

frequency of Model B-S becomes smaller and is similar to that of Model B-L after 8 seconds.

Figure 5 shows the time histories of the footing

and the ground surface displacements and Fig. 6

shows those of relative displacements between

the footing and the ground surface for both

models. The displacements were calculated by

56
the double integration of accelerometer recordings. The relative displacement, AD can be

defined by

AD = AB - AS (1)

in which AB = displacement of the footing and

AS = displacement of the ground surface. The

footing displacement is almost the same as the

ground surface displacement at first. A small

phase difference occurs between the ground

surface displacements and the footing after 8

seconds. This produces the relative displacement as shown in Fig. 6. The phase difference of

Model B-L is larger than that of Model B-S.

Thus, the relative displacement of Model B-L is larger than that of Model B-S as shown in Fig. 6.

3.2 Evaluation of Earth Pressure on Embedded

Foundation

Figure 7 is a schematic figure showing the seismic earth pressure acting on an embedded footing. The total earth pressure acting on the footing, P can be expressed by

P = Pp -Pa =Q-F (2)

in which Pp = earth pressure at passive side, Pa =

earth pressure at active side, Q = the sum of

shear forces at the pile heads and F = the inertial

force of the building, that is, the sum of the inertial forces of the super-structure and the footing. Q can be calculated by the differentiation of

the measured strain at the pile head.

In order to evaluate the accuracy of the shear

force at the pile heads, the shear force and the

inertial force of the building for Model A-L are compared in Fig. 8. The test set up of Model AL is similar to that of Model B-L except that the

footing was not embedded. The earth pressure

therefore does not act on the footing in Model AL. The shear force at the pile heads and the inertial force show a fairly good agreement, indicating that the total earth pressure can be evaluated by Eq. (2).

Figure 9 shows the time histories of the total

earth pressure on the footing for both models.

The total earth pressure recording contains

higher frequency components in Model B-S than

in Model B-L until 8 seconds. The tendency is similar to the acceleration of the super-structure

as shown in Fig. 2(a). From 8 to 20 seconds

when soil liquefaction is developing, the total

earth pressure is large and contains low frequency components for both models despite the

difference in the natural frequency of the superstructure. The total earth pressure gets small

after 20 seconds when the saturated sand liquefies completely.

To examine the effects of the inertial force and

the soil deformation on the total earth pressure,

fourier spectra of the total earth pressure, the

inertial force of the building and the ground displacement are shown in Fig. 10. The predominant frequencies of the total earth pressure are 3.6 Hz and 1.2 Hz for Models B-S and B-L, re- spectively until 8 seconds. The frequency agrees

with the predominant frequency of the inertial

force. This indicates that the total earth pressure

is strongly correlated to the response of the su- per-structure before soil liquefaction. The predominant frequency of the total earth pressure

for both models is 0.6 Hz, which corresponds to the predominant frequency of the soil displacement from 12 to 28 seconds. This suggests that

the total earth pressure is induced mainly by the

soil deformation after liquefaction.

3.3 Relation between Earth Pressure and Relative Displacement

To clarify the mechanism of the total earth pres- sure acting on the footing, the relation between

the total earth pressure and the relative displacement (AD) for Models B-S and B-L is shown in Fig. 11. The relation between the

relative displacement and the total earth pressure

is linear for both models until about 8 seconds.

It becomes nonlinear with the development of

the pore water pressure from 8 to 20 seconds.

The nonlinearity in behavior is significant in

Model B-L. The total earth pressure is seen to

degrade with cyclic loading from 20 to 30 sec- onds.

3.4 Relation between Earth Pressure and Inertial

Force of Building

57
In general, the phase difference between inertial

and kinematic forces can be classified into the

following two types in which Fb and Fg are natural frequencies of the building and the

ground, respectively (Nishimura et al, 1997):

1) If Fb > Fg

, inertial and kinematic forces tend to be in phase.

2) If Fb < Fg

, inertial and kinematic forces tend to be out of phase by 180 degrees.

The phase differences above are based on an

assumption that the total earth pressure and the

ground displacement should be in phase. The

total earth pressure acting on the footing, however, depends on the relative displacement.

Taking into account the displacement of the soil

and the footing, the phase can be classified into

four types as shown in Fig. 12.

(a) If Fb > Fg and AS > AB, the total earth pres- sure tends to be in phase with the inertial force.

Thus, the shear force at the pile heads corre- sponds to the sum of the total earth pressure and

the inertial force of the building.

(b) If Fb < Fg and AS > AB, the total earth pres- sure tends to be out of phase by 1 80 degrees with

the inertial force. Thus, the shear force at the

pile heads corresponds to the difference between

the total earth pressure and the inertial force of

the building.

(c) If Fb > Fg and AS < AB, the total earth pres- sure tends to be out of phase by 180 degrees with

the inertial force. Thus, the shear force at the

pile heads corresponds to the difference between

the total earth pressure and the inertial force of

the building.

(d) If Fb < Fg and AS < AB, the total earth pres- sure tends to be in phase with the inertial force.

Thus, the shear force at the pile heads corre- sponds to the sum of the total earth pressure and

the inertial force of the building.

To investigate the phase difference during soil

liquefaction, the relation between the total earth

pressure and the inertial force of the building for

the two tests is shown in Fig. 13. The data fallen

in the first and third quadrants show that the total

earth pressure tends to be in phase with the inertial force, while those in the second and forth

quadrants show that the total earth pressure

tends to be out of phase by 180 degrees with the

inertial force. A gray (blue) line in the figure

shows that AS is smaller than AB, while a black

line shows that AS is larger than AB.

Most of the data for both models fall in the sec- ond and forth quadrants until 8 seconds. This

indicates that the total earth pressure tends to be

out of phase by 180 degrees with the inertial

force. In case of Model B-S, a gray (blue) line is dominant, indicating that AS is smaller than AB

when the total earth pressure reaches its peak.

Considering that Fb is higher than Fg in this period, the total earth pressure tends to be out of

phase with the inertial force as shown in Fig. 12

(c). In case of Model B-L, on the contrary, a

black line is dominant, indicating that AS tends

to be larger than AB. Fb is smaller than Fg in this

period. Thus, the inertial force and the total

earth pressure tend to be out of phase as shown

in Fig. 12(b). The value of the total earth pres- sure in this period is about 60-70 percent of the

inertial force for both models. This indicates

that the non-liquefied crust tends to counteract

the inertial force transmitted from the building to

the pile heads. Therefore, the bending moment

at the pile heads is very small until 8 seconds as shown in Fig. 3.

Most data for both models fall in the first and

third quadrants from 8 to 20 seconds, indicating

that the total earth pressure tends to be in phase

with the inertial force. A black line is dominant,

indicating that AS tends to be larger than AB

when the total earth pressure reaches its peak.

Considering that Fb is higher than Fg due to liquefaction in both models, the total earth pressure

tends to be in phase with the inertial force as shown in Fig. 12(a). This indicates that the shear

force at the pile heads corresponds to the sum of

the total earth pressure and the inertial force of

the building. Therefore, the bending moment at the pile heads increase rapidly at 8 seconds as shown in Fig. 3. The tendency above is significant in Model B-L and is ambiguous in both

models from 20 to 30 seconds.

3.5 Effects of Natural Frequency of Building on

Earth Pressure

The peak value of the total earth pressure in

58
Model B-L is larger than that in Model B-S from

8 to 20 seconds as shown in Fig. 13. Taking into

account the similar soil response for both models

(Fig. 2,4), the difference may be caused by the

natural frequency of the super-structure. To

clarify the effects of the natural frequency of the

super-structure on the total earth pressure, the

relation between the acceleration of the footing

and the super-structure for the two tests is shown

in Fig. 14. The acceleration of the footing and

the super-structure tend to be in phase in Model

B-S. This indicates that the inertial force of the

super-structure does not interrupt the response of

the footing. The footing can move with the

ground and the relative displacement tends to be

small. In case of Model B-L, in contrast, the

acceleration of the super-structure tends to be

out of phase with that of the footing. This indicates that the inertial force of the building may

interrupt the response of the footing that tends to

move with the ground. As a result, the relative

displacement is large. This is the reason that, the

peak value of the total earth pressure for Model

B-L is larger than that of Model B-S.

4. CONCLUSION

Shaking table tests are conducted using a largescale laminar shear box to investigate the effects

of non-liquefied crust overlying liquefied soil on

an embedded footing. The following conclusions are drawn:

(1) The total earth pressure before liquefaction is induced mainly by the inertial force of the

building, because the total earth pressure acting

on the footing tends to be out of phase by 180

degrees with the inertial force of the building.

The shear force at the pile heads corresponds to the difference between the total earth pressure

and the inertial force.

(2) The total earth pressure after liquefaction is induced mainly by the soil deformation, because

the total earth pressure acting on the footing

tends to be in phase with the inertial force of the

building. The shear force at the pile heads cor- responds to the sum of the total earth pressure

and the inertial force.

(3) The relation between the relative displacement and the total earth pressure is linear before

liquefaction. It becomes nonlinear with the development of the pore water pressure and the

total earth pressure decreases with cyclic loading

after liquefaction.

(4) The peak value of the total earth pressure for

the super-structure with a low natural frequency

is larger than that with a high natural frequency.

This is probably because that the inertial force of

the super-structure with a low natural frequency

may interrupt the response of the footing that

tends to move with the ground.

ACKNOWLEDGMENTS

This test was jointly conducted by NIED, STA,

Tokyo Institute of Technology, Kajima Corp.,

Taisei Corp., Takenaka Corp., Nippon Steel

Corp. and Tokyo Soil Research Corp. The contribution and support of the members for this

work are appreciated.

REFERENCES

Fujii, S., Isemoto, N., Satou, Y., Kaneko, O.,

Funahara, H., Arai, T. and Tokimatsu, K.

[1998]. Investigation and analysis of a pile

foundation damaged by liquefaction during the

1995 Hyogoken-Nambu earthquake, Soils and

Foundations, Special Issue on Geotechnical Aspects of the 1995 Hyogoken Nambu Earthquake,

No. 2, pp. 179-192.

Nishimura, A., Murono, Y. and Nagatsuma, S.

[1997]. Experimental study on the seismic design method for pile foundation in the soft

ground (Part 1-3), Pros., 32nd Japan National

Conf. on Geotech. Engrg., Vol. 1, pp. 961-966,

(in Japanese)

Oh-Oka, H., Fukui, M., Hatanaka, M., Ohara, J. and Honda, S. [1998]. Permanent deformation

of steel pipe piles penetraiting compacted fill at wharf on port Island, Soils and Foundations,

Special Issue on Geotechnical Aspects of the

1995 Hyogoken Nambu Earthquake, No. 2, pp.

147-162.

Sato, M., Shamoto, Y. and Zhang, J. -M. [1995].

59
Soil pile-structure during liquefaction on centrifuge, Proc, 3rd Inter. Conf. on Rec. Adv. in

Geotech. Earthq. Engrg & Soil Dyn., 1, pp. 135-

142.

Tamura, S., Suzuki Y., Tsuchiya T., Fujii S. and

Kagawa T.[2000]. Dynamic response and failure

mechanisms of a pile foundation during soil liquefaction by shaking table test with a large-scale

laminar shear box, Proc, of 12th World Conf. on

Earthq. Engrg., Reference No. 903.

Tokimatsu, K. and Asaka, Y. [1998]. Effects of

liquefaction-induced ground displacements on

pile performance in the 1995 Hyogoken-Nambu

Earthquake, Soils and Foundations, Special Is- sue of Geotechnical Aspects on the 1995

Hyogoken Nambu Earthquake, No. 2, pp. 163-

177.
accelerometer (Horizontal) o pore pressure

rji accelerometer (Cross) • strain gage

O accelerometer (Vertical) A rotary encoder

> horizontal displacement

V vertical displacement

(Unit: mm)

Fig. 1. Test setup(Model B-S)

100 e L (a) B-S

-100

100

. 0 A

-100

Cb) B-L Ground Surface

•VMM/

Footing

0 5 10 15 20 25 30 Time (s) Fig. 5. Time histories offooting and ground displacement

~ 40

~ 0

"-40 — 40

(a) B-S

(b) B-L

15 Time (s)20

Fig. 6. Time histories of relative displacement

between footing and ground surface

Inertial Force of Building.F

Active Side !H1

Passive Side

Shear Force at Pile Head.O

Fig. 7. Schematic figure showing seismic earth pressure

acting on footing

61
200

-200

200

-200

400

= 0 o

-400

1 1 1 I | 1 1 I 1 | I I I 1 L|

i k - I k

,

. . . .

la) Super-structure , *

,

B-S

B-L

(c) input Acceleration

= 0

1

1 1 1 B-S

(bl Ground surface

:

. 1 .... 1 . . . . 1 . . . .

5 10 IS 20 25 30 Tile Is)

Fig. 2. Time histories ofacceleration

1 1 1 1

1

1 1 . 1 11

-^A/U

, '-^Vs 1

'

Pile Head

i : i i 15 10 15 20 25 Tiie Is)

Fig. 3. Time history ofbending moment at pile head

15 Tue Is)

Inertia! Force of Building Shear Force at PBe, Head,

Fig.8. Time history ofinertialforce ofbuilding and

shearforce at pile head

-40

~r ~r

(a) B-S

.... i .... i ... i .... i ..

I I . (b) B-Li

, 1 1 ,

1 1 . ,

,

i . . . . 15 Time (s)

20 25

Fig. 9. Time histories oftotal earth pressure acting onfooting

—

i

•—>—1 1

i

i—

i

—1 i

_L_

a.

ui n 15 Time (s)

^\ifa v t=16s \* \„.

v

-t=! t=12s t=20s

J I I 1_

1 1 1 1 1

1 - lb) G.L-41

I I I I

B-S

. B-L

'•'<

-ff^V"i 1—i—i_ .... i ... .

'0 10 20 30 40 50 600 10 20 30 40 50 60

Excess Pore Water Pressure(kPa) Excess Pore Water Pressure(kPa)

FigA.Time histories and vertical distribution of

excess pore water pressure

(c) Ground displacement (B-S)

,t=12-20s

\

\ t=20-28s

if V'"1

I

\* '. t=0-8s

I 2 3 4 5 6 T

Frequency (Hz)

(f) Ground displacement (B-L)

,'.t=1Z-20s

2f- i j t=20-28s

'I %\ .1=0-8.

1 2 3 4 5 Frequency (Hz)

Fig. 10. Fourier spectra of earth pressure, inertialforce

and soil displacement

62
10

S 0 -

-10

(a) t=0-8s (B- S)

0.5 0 tt 5 -0.5 o a 5 Relative Displacement (cm) Relative Displacement (cm)

X 0

-30

lc)t=20-30s IB-S) j

-30

lf)t=20-30s IB

(J

Relative Displacement (cm)

-5 0 5 Relative Displacement (cm)

Fig. 11. Relation between relative displacement

and total earth pressure

30

2 0

18

S o

la)t=0-8s (B-S) mm ! ip^l

lb) t=8-20s B-S)

•

(f) t=20-30s (B-L)

"30 0 30 -30 0 30

Inertial Force of Structure IkN) Inertial Force of Structure (kN)

Fig. 13. Relation between inertialforce and

total earth pressure

Inertial force of building

Footing displacement

AS AB

ft II II

Ground displacement

/

AS

h-t>

Earth Pressure

I I I

AS

(a) Fb > Fg & AS > AB

u'

' AB AS

I

(c) Fb > Fg & AS < AB

F

AS AB

ii ii

AS

|-t>

I

(b) Fb < Ff & AS > AB

F

AS AB I h-fr' T

AS

! nr" i

(d) F„<F, & AS<AB

Fig. 12. Schematicfigure showing relation between

inertialforce and total earth pressure

500 0 , . 500

Acc. of Footing (cm/f)

0 , ,500

Acc. of Footing (cm/s )

Fig. 14. Relation between acceleration offooting

and super-structure

63

SESSION 2

NEXT-GENERATION BUILDING

AND INFRASTRUCTURE

SYSTEMS

Project on 3-D Full-Scale Earthquake Testing Facility

(The Second Report)

by

Keiichi Ohtani1

, Nobuyuki Ogawa2

, Tsuneo Katayama3

, Heki Shibata4

ABSTRACT

The Hanshin-Awaji Earthquake (January 17,

1995) clearly demonstrated that the occurrence of

very strong ground motion in the area near to the

seismic fault is capable of causing severe

structural damage beyond general estimation. It has emphasized the importance of earthquake

engineering research into why and how structures

collapse in real earthquake conditions.

Considering the lessons learnt from recent

earthquake disasters, NIED plan to construct a "3-D Full-Scale Earthquake Testing Facility

("E-Defense" is the nickname of this facility)",

which will be able to simulate the processes of

destruction of structures under the condition of

real strong earthquake motions.

The basic performances of "E-Defense" are maximum lording capacity 1,200 tons, maximum

velocity 200 cra/s and maximum displacement 2m

p-p for horizontal excitation and maximum

velocity 70 cm/s, maximum displacement lm p-p

for vertical excitation to realize destructive ground

motion. The construction work of "E-Defense"

has began at early 2000, five year after the

Hanshin-Awaji Earthquake and will be completed

at the beginning of 2005, ten years after that

Earthquake. Now, we are conducting the

construction works of the facility at the Miki-city, near Kobe-city, and the manufacturing of

actuators, oil-pressure supply system and other

major parts of shaking table at the Mitsubishi

Heavy Industry Co.

"E-Defense" is the very large scale and high

performance testing facility in the world.

Therefore, many researchers, which are belonging

not only Japanese but also worldwide

organizations, can use this facility for their

researches. "E-Defense" should be operated the

international common use. For the international

collaboration and the dissemination of research

results (including test data), Earthquake

Engineering Network ("EE-net") will also

construct until the completion of "E-Defense".

EE-net will connect, through a high performance

Internet, distributed major earthquake engineering

research organization.

We consider that the researchers together from

worldwide and research projects will determine

and evaluate by the International Committee.

We hope that "E-Defense" and EE-net will be

situated to one of the cooperative research

organization for the earthquake disaster mitigation

in the world.

KEYWORDS:

Failure mechanism of structures

Full Scale Testing

International Common Use

3-D Shaking Table

Network of Earthquake Research

' Project Director, 3-D Full-Scale Earthquake Testing Facility (E-Defense), National Research Institute

for Earth Science and Disaster Prevention (NIED), Tsukuba, Japan

2 Project Sub-Director, E-Defense, NIRD, Japan

3 President, NEED, Japan

4 Visiting Research Fellow, NEED, Japan

67
1. INTRODUCTION

The Hanshin-Awaji Earthquake

(Hyogoken-Nanbu Earthquake, January 17, 1995)

clearly demonstrated that the occurrence of very

strong ground motion in the area near to the

seismic fault is capable of causing severe

structural damage beyond general estimation.

The destructive earthquake occurred in the

worldwide in the recent years, such as Northridge

earthquake (1994), Umbria-Marche earthquake

(1997), Kocaeli earthquake (1999), Ji-ji

earthquake (1999), El Salvador earthquake (2000),

Gujarat earthquake (2001) and so on.

In order to reduce the hazards associated with

large earthquakes, it is essential to improve the

reliability of earthquake resistance estimations and

reinforcement methods in the construction of

urban and major structures. For this purpose,

failure mechanisms and collapse processes of

various kinds of full-scale structures must be

investigated. Many types of experimental

apparatus have been used for such investigations,

and some of them have as large a size as possible

to alleviate any difficulties arising from limitation

of the model>Considering the lessons learnt from

recent earthquake disasters, the National Research

Institute for Earth Science and Disaster Prevention

(NEED) planned to build a new three-dimensional,

full-scale, earthquake testing facility, which can

carry large-size soil and structure models and

reproduce the processes of structural failure.

This facility is expected to become a powerful

tool for international collaboration in earthquake

engineering research. It also requires

international cooperation to successfully complete

the facility and to use it effectively for

engineering purposes.

Following the technical developments and

surveys in earthquake engineering and related

fields, the NIED began the design and

construction of this new facility in the Japanese

fiscal year of 1998. This paper summarizes the

construction plan and Earthquake Engineering

Network (EE-Net), which is the tool for ensure of

the international collaboration and the

dissemination of research results.

2. E-DEFENSE (3-D FULL-SCALE

EARTHQUAKE TESTING FACILITY)

Based on the lessons learnt from Hanshin-Awaji

earthquake, the Minister of State for Science and

Technology was inquired to the Council for

Aeronautics, Electronics and Other Advanced

Technology, which is the one inquire organization

of the Minister, for the discussion of the effective

arrangement of research bases for earthquake

disaster mitigation at March 29,1996. The

Council was reported to the Minister at September

3, 1997.

The report was clearly pointed out the

arrangement of large-scale three-dimensional

earthquake simulator facility as the core facility of

research bases for earthquake disaster mitigation.

NIED initiated the project on the large-scale

three-dimensional earthquake simulator facility

just after the occurrence of Hanshin-Awaji

earthquake. The research and development for

core technology for this facility (E-Defense) was

started on 1995. The fundamental concepts of

this project based on the report by the council.

The E-Defense will construct as the core facility

of the research bases for earthquake disaster

mitigation. Therefore, we need to clear the

positions of the E-Defense.

1) Position of earthquake simulator for the main

element of development of the 'Time-Space

Domain Simulation System for Earthquake

Disaster.)

2) Position of the clearly understanding of failure

mechanism of structures.

3) Position of the response mechanism for the

request from major subject of earthquake

engineering.

The importance of promoting the strengthening

68
and rationalization of earthquake-proof structural

design is just one of the lessons from

Hanshin-Awaji earthquake. Because earthquake

vibrations involve three-dimensional movement, it

is necessary to set up a three-dimensional

earthquake simulator facility to accurately

reproduce earthquake motions. To perform tests on real-size objects or large-scale models of test

structures and foundations, it is desirable to have

the large-scale three-dimensional shaking table.

If large-scale 3 -dimensional shaking table is

available, tests could be performed to shed new

light on the mechanism of dynamic failure using

real-size structures. If a stage reached whereby

design based on such discovery can be performed,

this will contribute immensely to reducing

earthquake disaster.

The main specification of E-Defense is shown in

Table 1 . The actuator performance for horizontal

and vertical axes is shown in Fig. 1

.

3. CONSTRUCTION AND MANUFACTURING

OF E-DEFENSE

NIED have commenced the development work of

shaking mechanism with very large size of

hydraulic actuators in fiscal year 1995 and

completed performance tests successfully in 1998.

Following the above technical development and

surveys in earthquake engineering and related

fields, NEED have began the design and

construction of E-Defense in 1998.

Figure 2 shows the drawing bird eye view of

E-Defense. We will construct several buildings,

such as laboratory building, measurement and

control building, hydraulic oil unit building,

preparation building and so on. The

3-dimensional shaking table will be installed in

the laboratory building. Hydraulic oil will be

supplied to shaking table by pipelines via

underground culvert. The reaction foundation

(shaking table foundation) has weight of about

2GN (200,000 tonf) and set to the bedrock

directly.

The construction work has been began in 1998

and will be completed at the beginning of 2005.

The new facility will start to operate at the 10

years after the Hanshin-Awaji earthquake. The

E-Defense is constructed in "Miki Earthquake

Disaster memorial Park (tentative name)", which

is being constructed in Miki city, on the north of

Kobe city. The construction of shaking table

foundation was started at the construction site in

January 2000. Figure 3 shows the aerial

photograph of the site before the construction

work. Figure 4 shows the scene of the first

concrete casting for the foundation. The D51

(diameter 51 mm) reinforcing bars used for the

foundation, such as the foundation for Nuclear

Power Plant. Figure 5 is the recent construction

condition.

The manufacturing of testing equipments, such as

actuators, 3-dimensional link joint, oil power

pump unit, accumulator unit and so on, were

started in 1998. By the construction of

manufactured unit are limited some size by the

condition of transportation. The setup working

will be done at the site. Figure 6 shows the

set-upped actuator in the preparation building.

4. EARTHQUAKE ENGINEERING NETWORK

The report by the Council was strongly suggested

that E-Defense should be operated the

international common use. It is important to arrange the utilizing structures, equipment and

support section for outside users. To ensure the

international common use and disseminate the test

results, we will construct and install the

Earthquake Engineering Network (EE-Net).

The EE-Net has tow major functions: The one is

the connection tool between E-Defense in Miki

and the Super Computer in Tsukuba. The other

one is the connection tool between NIED and the

other organization, such as research institute,

university, private sector and so on. This

function is not only limited to domestic, but also

international manner.

69
We will install the tele-observation and

tele-discussion capabilities, but not install the

tele-operation function. Because, conducting of

shaking table test, especially failure test, has very

delicate and dangerous factors. Therefore, the

operation of shaking table will limit by the

shaking table administrator, who is the specialist

of operation. Figure 7 shows the schematic image

of EE-Net, and Table 2

Shows the security control system for different

kind of users.

5. CONCLUDING REMARKS

Based on the lessons learnt from Hanshin-Awaji

earthquake, we, NIED, need more research to

understand the failure mechanism of different

kind of structures during earthquake. For this

research needs, we began the construction project

of E-Defense (3-D Full-Scale Earthquake Testing

Facility) and EE-Net (Earthquake Engineering

Network). After completion, these tools will be

perfectly opened to international use. We strongly hope that these tools are contributed

to the dramatic progress of the earthquake

engineering research, especially the understanding

of structural failure mechanism, the progress of

the earthquake resistant design of structures and

the evaluation/reevaluation of structural

performance during earthquake, by the

coordination and collaboration research works in

the worldwide bases.

REFERNCES:

Ohtani, K, Katayama, T, Shibata, H. (1998) "Project on 3-D Full-Scale Earthquake Testing Facility - First

Report-", 30UJNR

Ogawa, N, Ohtani, K, Katayama, T, Shibata, H. (1999)

"World's Largest Shaking Table Takes Shapes in

Japan - A Summary of Construction Plan and

Technical Development-", SmiRT 16

Ohtani. K, Ogawa, N, Katayama, T, Shibata, H,

Nakagawa, O, Ohtomo, T. (2001) "World's Largest

Shaking Table Takes Shapes in Japan (The 2

nd

Report)",S/m7?7'77

Ohtani, K. (2001) 'Tools for Seismic Safety Evaluation

of Structures", NEA/CSNI/R(2001)13/VOL3

Ohtani, K, Ogawa, N, Katayama, T, Shibata, H. (2002)

"3-D Full-Scale Earthquake Testing Facility and

Earthquake Engineering Network", 3WCSC

Payload 12MN(1200tonf)

Size 20m X 15m

Driving Type

Accumula

Electro-Hydraul

tor Charge

c Servo Control Shaking Direction X-Y - Horizontal Z-Vertical

Maximum Acceleration

(at Maximum Loading) >900cm/s2 >1,500cm/s2 Maximum Velocity 200cm/s 70cm/s

Maximum Displacement ±100cm ±50cm

Moment verturning Momn< Yawing Moment

... ........ . 150MN-m 40MN-m

Table 1 . Main Specification of E-Defense

0.1 1.0 10.0 50.0 Fraquency(Hz)

Fig. 1 . Limit Performance of E-Defense

70
71

Fig. 6 Set-upped Horizontal Actuator

Fig. 7 Schematic Images of EE-Net

Table 2 Security Control for Users

Shaking

taole

Con:ro:

, Program

Execute mathematic a; TV

Conference Access tc

operation jploaa simulation Pa'rticipau.o4 ;est data

Shaking table administrator o o O o o

Shaking table User

(at Miki) X o O o o

Mathematical simulator administrator X o o o o

Mathematical simulator user

(at Miki) X o o o 0

Mathematical simulator user (atTSUKUBA)

X o o o 0

Network administrator X o o

Network User

(at Miki)

X o o

Network User

(at TSUKUBA)

X o o

RABnet User X o o o o

US NEES User X o o o o

EU JRC User X o o o 6

Internet User X X X X Q (Smtefl access)

O: Accessible X: Inaccessible —: Accessible depending on user skill

73

I

Smart Damping Technologies for Dynamic Hazard Mitigation

by

B.F. Spencer Jr.

1

ABSTRACT

In recent years, considerable attention has been

paid to research and development of structural

control devices, with particular emphasis on alleviation of wind and seismic response of buildings and bridges. In both areas, serious efforts

have been undertaken to develop the structural

control concept into a workable technology. To

date, full-scale active and hybrid control systems

have been designed and installed in approximately 40 commercial buildings and 1 5 bridges (during construction). Yet the engineering

community is reluctant to fully embrace this new

technology. Demonstrated cost-effectiveness and

reliability are key considerations for acceptance

and successful implementation of structural control. Because of their low power requirements

and fail-safe character, smart damping strategies

appear quite attractive in this regard. The focus

of this paper will be to review a number of smart

damping approaches that have been proposed

and implemented in full-scale structures.

1.0 INTRODUCTION

Passive supplemental damping strategies, including base isolation systems, viscoelastic dampers

and tuned mass dampers, are well understood

and are accepted by the engineering community

as a means for mitigating the effects of dynamic

loadings such as strong earthquakes and high

winds. However, these passive-device methods

are unable to adapt to structural changes and to varying usage patterns and loading conditions.

For more than two decades, researchers have investigated the possibility of using active control

methods to improve upon passive approaches to reduce structural responses (Soong 1990; Spencer and Sain 1997; Housner et al. 1997; Soong

and Spencer 2002). The first full-scale application of active control to a building was accomplished by the Kajima Corporation in 1989

(Kobori 1994). The Kyobashi Seiwa building is an 11-story (33.1 m) building with a total floor area of 423 m . An active mass driver (AMD)

system was installed, consisting of two AMDs —

the primary AMD is used for transverse motion

and has a mass of 4 tons, while the secondary

AMD has a mass of 1 ton and is employed to re- duce torsional motion. The role of the active system is to reduce building vibration under strong

winds and moderate earthquake excitations and

consequently to increase the comfort of occupants of the building. Since that time, active/hybrid structural control has been successfully

applied in approximately 40 commercial buildings and 15 bridges (during construction).

Although extensive analytical and experimental

structural control research has been conducted in

both the U.S. and Japan in the last decade, with

the exception of one experimental system installed on a bridge in Oklahoma (discussed later

in this paper), none of these full-scale active control installations are located the U.S. Many possible reasons can be cited for this disparity. For

example, the civil engineering profession and

construction industry in the U.S. are conservative

and generally reluctant to apply new technologies. The absence of verified and consensus-approved analysis, design and testing procedures

represent additional impediments to the application of this technology. However, more notable is the lack of research and development expenditures by the U.S. construction industry. This situation is in sharp contrast to the Japanese

construction industry, which invests heavily in

the development and implementation of new

Dept. of Civil Engineering and Geological Sciences, University of Notre Dame, 1 56 Fitzpatrick Hall,

Notre Dame, IN 46556 USA.

75
technologies. Yet even in Japan, few new

projects for implementation of active control systems are being initiated. This situation is partly

due to the modest number of tall buildings and

long-span bridges being planned for the near future and partly due to a number of serious challenges that remain before active control can gain

general acceptance by the engineering and construction professions at large. These challenges

include: (i) reducing capital cost and maintenance, (ii) eliminating reliance on external power, (iii) increasing system reliability and

robustness, and (iv) gaining acceptance of nontraditional technology.

Despite the impediments that exist to wider application of control to civil engineering structures, the future appears quite bright. Smart

damping (also known as semiactive control)

strategies are particularly promising in addressing many of the challenges to this technology, offering the reliability of passive devices, yet

maintaining the versatility and adaptability of

fully active systems. The remainder of this paper

discusses several smart damping strategies that

have recently been proposed, both in the U.S.

and in Japan, for control of civil engineering

structures, as well as several applications of this

technology. Such systems may facilitate near- term acceptance of control technology by practitioners as an important means for mitigating dynamic hazards.

2.0 SMART DAMPING DEVICES

Smart damping devices have received a great

deal of attention in recent years because they offer the adaptability of active control devices

without requiring the associated large power

sources. In fact, many can operate on battery

power, which is critical during seismic events

when the main power source to the structure may

fail. According to presently accepted definitions, a smart damping device is one which cannot inject mechanical energy into the controlled structural system (i.e., including the structure and the

control device), but has properties that can be

controlled to optimally reduce the responses of

the system. Therefore, in contrast to active control devices, smart damping devices do not have

the potential to destabilize (in the bounded input/

bounded output sense) the structural system.

Studies have shown that appropriately implemented smart damping systems perform significantly better than passive devices and have the

potential to achieve, or even surpass, the performance of fully active systems, thus allowing for

the possibility of effective response reduction

during a wide array of dynamic loading conditions (Dyke et al. 1998; Spencer et al. 2000). Examples of such devices include variable-orifice

fluid dampers, controllable friction devices, variable stiffness devices, adjustable tuned liquid

dampers, and controllable fluid dampers (Spencer and Sain 1997). Of these classes of smart

dampers, two have already been implemented in

full-scale structures and will be discussed in the

subsequent sections.

2.1 variable-orifice Dampers

One means of achieving a smart damping device

is to use a controllable, electromechanical, variable-orifice valve to alter the resistance to flow

of a conventional hydraulic fluid damper. A

schematic of such a device is given in Fig. 1, which typically operates on approximately 50

watts of power.

Figure 1. Schematic of Variable-Orifice Damper.

Sack and Patten (1993) conducted experiments

in which a hydraulic actuator with a controllable

orifice was implemented in a single-lane model

bridge to dissipate the energy induced by vehicle

traffic. These studies were followed by a fullscale experiment conducted on a bridge on interstate highway 1-35 to demonstrate this technology (Patten, 1998, 1999; Kuehn et al., 1999)

shown in Figs. 2-3. Figure 4 shows the effective- ness of the SAVA system. This experiment constitutes the first full-scale implementation of

structural control in the US.

76
Peak Stress on the Middle Cross Section, Heavy Truck /Left Lane

Figure 2. First Full-Scale Implementation of

Smart Damping in the US.

s

? -2000

| -1500

itI -2500

r»— ei—.—s_,

Controlle d

\

.

V

\

1 \ \

\ \ \ \ \ \

No Contra \ \ \ 1 \

c i

2 3 4 Girder #. Eastern to Western

Peak Stress on the Middle Cross Section, Heavy Truck /Right Lane

c Conor Ued //

/ J

8

J

r CV

Ho Control

12 3 4 Girder #, Eastern to Western

Figure 4. Comparison of Peak Stresses for Heavy

Trucks.

Figure 3. SAVA-II variable-orifice Damper.

Conceived as a variable-stiffness device, Kobori

et al. (1993) and Kamagata and Kobori (1994)

implemented a full-scale variable-orifice damper in a semiactive variable-stiffness system

(SAVS) to investigate semiactive control at the

Kajima Technical Research Institute. The overall

system is shown in Fig. 5 where SAVS devices were installed on both sides of the structure in

the longitudinal direction. The results of these

analytical and experimental studies indicate that

this device is effective in reducing structural re- sponses.

More recently, a smart damping system was installed in the Kajima Shizuoka Building in Shizuoka, Japan. As seen in Fig. 6, semiactive

hydraulic dampers are installed inside the walls on both sides of the building to enable it to be

used as a disaster relief base in post-earthquake Figure 5. SAVS System Configuration.

77
situations (Kobori, 1998; Kurata et al., 1999).

Each damper contains a flow control valve, a check valve and an accumulator, and can develop

a maximum damping force of 1000 kN (see Fig.

7). Figure 8 shows a sample of the response analysis results based on one of the selected control

schemes and several earthquake input motions

with the scaled maximum velocity of 50 cm/sec,

together with a simulated Tokai wave. Both story

shear forces and story drifts are seen to be greatly

reduced with control activated. In the case of the

shear forces, they are confined within their elastic-limit values (indicated by E-limit) while,

without control, they would enter the plastic

range.

I—«— EL50 - TFSO

t I HA50

\l I—o-- Tokai ^ ELSO ! \ -A- TFSO

V. 1 • D -HA50 : V l-o-- Tokai j

! \

^ v

-

f \

s «

1

\ /

0 2000 4000 6000 8000(kN) 0 2 4 6 8 I0(cm)

Maximum .tarlora Maximum Mory drift

(a) w/ control (b) w/o control

Figure 8. Maximum Responses (El Centra, Taft

and Hachinohe Waves with 50 cm/sec. and Assumed Tokai Waves).

Figure 6. Kajima Shizuoka Building Configured

with Semiactive Hydraulic Dampers.

Figure 7. Shizuoka Building variable-orifice

Damper.

The use of the variable-orifice damper has blossomed in Japan. In the Tokyo Siodome area, 4

new buildings are currently under construction

(Kobori 2002). One of these structures is the Kajima K-Building, 38-story building with 88 variable-orifice dampers and 2 hybrid mass dampers.

In another area of Tokyo, the Kajima R-Building,

a 54-story building with 356 variable-orifice

dampers and 192 passive dampers distributed

throughout, is under construction. When these

projects are completed, a total of more than 700

variable-orifice dampers will be installed in

building structures in Japan.

2.2 Controllable Fluid Dampers

In comparison with variable-orifice damper sys- tems, another class of relatively new smart

damping devices uses controllable fluids, schematically shown in Fig. 9. In comparison with

smart damping systems described above, an advantage of controllable fluid devices is that they

contain no moving parts other than the piston,

which makes them simple and potentially very

reliable.

Electromagnetic Choke

MR Fluid £x^l

Figure 9. Schematic of MR Damper.

78
(a)

Thermal Expansion

Accumulator

Figure 10. (a) Schematic of 20-ton MR fluid damper; (b) Experimental setup.

The essential characteristics of controllable fluids is their ability to reversibly change from a

free-flowing, linear viscous fluid to a semi-solid

with a controllable yield strength in milliseconds

when exposed to an electric (for electrorheological (ER) fluids) or magnetic (for magnetorheological (MR) fluids) field.

In the case of magnetorheological fluids, they

typically consist of micron-sized, magnetically

polarizable particles dispersed in a carrier medium such as mineral or silicone oil. When a magnetic field is applied to the fluid, particle chains

form, and the fluid becomes a semi-solid and exhibits viscoplastic behavior. Transition to Theological equilibrium can be achieved in a few

milliseconds, allowing construction of devices

with high bandwidth. Additionally, Carlson and

Weiss (1994) indicated that high yield stress of a magnetorheological fluid can be achieved and

that magnetorheological fluids can operate at temperatures from -40°C to 150°C with only

slight variations in the yield stress. Moreover,

magnetorheological fluids are not sensitive to impurities such as are commonly encountered

during manufacturing and usage, and little particle/carrier fluid separation takes place in magnetorheological fluids under common flow

conditions. Further, a wider choice of additives

(surfactants, dispersants, friction modifiers, anti- wear agents, etc.) can generally be used with

magnetorheological fluids to enhance stability,

seal life, bearing life, and so on, since electrochemistry does not affect the magnetopolarization mechanism. The magnetorheological fluid can be readily controlled with a low voltage (e.g.,

12-24 V), current-driven power supply outputting only 1-2 amps.

Carlson and Spencer (1996) and Spencer et al. (1997, 1998) and Yang et al. (2002) report on the

design of a full-scale, 20-ton magnetorheological

damper (see Fig. 10-11) showing that this technology is scalable to devices appropriate for civil

engineering applications. At design velocities,

the dynamic range of forces produced by this device is over 10 (see Fig. 11), and the total power

required by the device is only 20-50 W.

1 1 1 1 1

Max Magnetic Field (2A)

No Magnetic Field (OA) L ) \ -T

, -y j

0

1 ' ' 1 ' ' 1 -6 -4 -2 0 2 4 6

Displacement (cm)

Figure 1 1 . Measured force-displacement loops at 5.4 cm/sec.

79
Recently, Sunakoda, et al. (2001) have also presented encouraging results regarding design,

construction, and commercial manufacturing of

large scale MR dampers.

In 2001, the first full-scale implementation of

MR dampers for civil engineering applications was achieved. The Nihon-Kagaku-Miraikan, the

Tokyo National Museum of Emerging Science

and Innovation shown in Fig. 12, has two 30-ton,

MR Fluid dampers installed between the 3rd and

5th floors. The dampers were built by Sanwa

Tekki using the Lord Corporation MR fluid.

Currently being retrofitted with stay-cable dampers, the Dongting Lake Bridge in Hunan, China

will constitute the first full-scale implementation

of MR dampers for bridge structures. Long steel

cables, such as are used in cable-stayed bridges

and other structures, are prone to vibration in- duced by the structure to which they are connected and by weather conditions, particularly wind

Figure 12. Nihon-Kagaku-Miraikan, Tokyo National Museum of Emerging Science and Innovation.

Figure 13. Dongting Lake Bridge, Hunan, China.

combined with rain, that may cause cable galloping. The extremely low damping inherent in such

cables, typically on the order of a fraction of a

percent, is insufficient to eliminate this vibration,

causing reduced cable and connection life due to

fatigue and/or breakdown of corrosion protection. Two Lord SD-1005 (www.rheonetic.com)

MR dampers are being installed on each cable to mitigate cable vibration.

3.0 CONCLUSIONS

Although in their infancy, control strategies

based on smart damping devices appear to combine the best features of both passive and active

control systems and to offer a viable means of

protecting civil engineering structural systems

against earthquake and wind loading. In particular, they provide the reliability and fail-safe character of passive devices, yet possess the

adaptability of fully active devices. Because of

their mechanical simplicity, low power requirements and high force capacity, magnetorheological (MR) dampers constitute a class of smart

damping devices that meshes well with the demands and constraints of civil infrastructure applications and will likely see increased interest

from the engineering community.

ACKNOWLEDGMENT

The author gratefully acknowledges the support

of this research by the National Science Foundation under grant CMS 99-00234 (Dr. S.C. Liu,

Program Director) and the Lord Corporation.

80
REFERENCES

Carlson, J.D. and Weiss, K.D. "A Growing Attraction to Magnetic Fluids," Machine Design,

August, pp. 61-64, 1994.

Dyke, S.J., Spencer Jr., B.F. Sain, M.K. and Carlson, J.D. "An Experimental Study of MR Dampers for Seismic Protection," Smart Materials and

Structures, Vol. 7, pp. 693-703, 1998.

Housner, G.W., Bergman, L.A., Caughey, T.K.,

Chassiakos, A.G., Claus, R.O., Masri, S.F., Skelton, R.E., Soong, T.T., Spencer, B.F. and Yao,

J.T.P. "Structural Control: Past, Present and Future," Journal ofEngineering Mechanics, ASCE,

Vol. 123, No. 9, pp. 897-971, 1997.

Kamagata, S. and Kobori, T. "Autonomous

Adaptive Control of Active Variable Stiffness

Systems for Seismic Ground Motion", Proceedings of the 1

st World Conference on Structural

Control, pp. TA4:33-42, Pasadena, CA, 1994.

Kobori, T. "Future Direction on Research and

Development of Seismic-Response-Controlled

Structure," Proceedings of the 1st World Confer- ence on Structural Control, Panel: 19-31 (1994).

Kobori, T. "Mission and Perspective Towards

Future Structural Control Research", Proceedings of the 2

nd World Conference on Structural

Control, 1, pp. 25-34, Kyoto, Japan, 1998.

Kobori, T. "Past, Present and Future in Seismic

Response Control in Civil Engineering Structures," Proceedings of the 3rd World Conference on Structural Control, 2002.

Kobori, T., Takahashi, M., Nasu, T., Niwa, N.

and Ogasawara, K. "Seismic response controlled

structure with active variable stiffness system",

Earthquake Engineering and Structural Dynamics, 22, pp. 925-941, 1993.

Kuehn, J., Song, G. and Sun, J. "Experimental

Verification of a NON-Protruding Intelligent

Stiffener for Bridges (ISB)", Proc. Int. PostSMiRT Conf Seminar on Seismic Isolation, Passive Energy, Dissipation and Active Control of

Vibrations of Structures, Cheju, Korea, August

23-25, 1999.

Kurata, N., Kobori, T., Takahashi, M., Niwa, N.

and Midorikawa, H. "Actual Seismic Response

Controlled Building with Semi-active Damper

System," Earthquake Engineering and Structural Dynamics, Vol. 28, ppl427-1447, 1999.

Patten, W.N. "The 1-35 Walnut Creek Bridge: An

Intelligent Highway Bridge Via Semi-active

Structural Control", Proceedings of the 2

nd

World Conference on Structural Control, Vol. 1,

pp. 427-436, Kyoto, Japan, 1998.

Patten, W., Sun, J., Li, G., Kuehn, J. and Song,

G. "Field Test of an Intelligent Stiffener for

Bridges at the 1-35 Walnut Creek Bridge". Earthquake Engineering and Structural Dynamics,

Vol. 28, No. 2, pp. 109-126, 1999.

Sack, R.L. and Patten, W. "Semiactive Hydraulic

Structural Control", Proceedings of the International Workshop on Structural Control, USC

Publication No. CE-9311, pp. 417-431, Honolulu, Hawaii, 1993.

Sodeyama, H. Sunakoda, K, Suzuki, K, Carlson, J.D. and Spencer Jr., B.F. "Development of

Large Capacity Semi-Active Vibration Control

Device using Magneto-Rheological Fluid," PVPVol. 428-2, Seismic Engineering-Volume 2, pp.

109-114,2001.

Soong, T.T. Active Structural Control: Theory

and Practice, Longman Scientific and Technical,

Essex, England (1990).

Soong, T.T. and Spencer Jr., B.F. "Supplementary Energy Dissipation: State-of-the-Art and

State-of-the-Practice," Engineering Structures,

Vol. 24, pp. 243-259, 2002.

Spencer Jr., B.F., Carlson, J.D., Sain, M.K. and

Yang, G. "On the Current Status of Magnetorheological Dampers: Seismic Protection of

Full-scale Structures", Proceedings of the Ameri- can Control Conference, pp. 458-462, Albuquerque, NM, 1997.

81
Spencer Jr., B.F. and Sain, M.K. "Controlling

Buildings: A New Frontier in Feedback," IEEE

Control Systems Magazine: Special Issue on Emerging Technologies (Tariq Samad, Guest

Ed.), Vol. 17, No. 6, pp. 19-35, 1997.

Spencer Jr., B.F., Yang, G., Carlson, J.D. and

Sain, M.K. "Smart dampers for seismic protection of structures: A full-scale study", Proceedings of the 2

nd World Conference on Structural

Control, 1, pp. 417-426, Kyoto, Japan, 1998.

Spencer Jr., B.F, Johnson, E.A. and Ramallo, J. C. "'Smart' Isolation for Seismic Control," JSME

International Journal: Special Issue on Frontiers

of Motion and Vibration Control, Series C, Vol.

43, No. 3, pp. 704-711,2000.

Spencer Jr., B.F, Yang, G., Carlson, J.D. and

Sain, M.K. "Large-Scale MR Fluid Dampers:

Modeling and Dynamic Performance Considerations," Engineering Structures, Vol. 24, pp. 309-323, 2002.

82
Research and Development Programs

on Timber Structures in Japan

Hiroshi Isoda

1

, Hisashi Okada2

, Naohito Kawai3 and Nobuyoshi Yamaguchi1

ABSTRACT

Recently there have been many projects

pertaining to timber structures, mainly dealing

with seismic behavior. This paper outlines two

main projects that currently are being

promoted by several committees within BRI

and external to BRI, which report to the BRI

project team. They are also being developed

collaboratively with public corporations, and

international institutes. The Final target of one

project is to develop high-performance hybrid

timber members and hybrid timber structures,

which consist of timber and other materials, to

develop the performance evaluation methods

for these hybrid timber structures, and to

develop the design method for some typical

hybrid timber structures. The other target is to

develop the seismic rehabilitation system for

existing inferior, low-rise wooden buildings.

These projects have several goals including

developing reliable and economical ways of

improving performance of residential wooden

houses as well as minimizing the

environmental costs in the construction of

new homes. The effective utilization of timber

is urged from the point of preservation of

natural resources and global environment also

in Japan.

KEYWORDS: timber structures, advanced

hybrid timber members and structures,

seismic performance, seismic rehabilitation,

fireproof performance

1. INTRODUCTION

Recently Building Research Institute is

promoting research and development

programs on timber structures as one of

important tasks. The research motivation is

that timber structures represent one of

society's largest investments in the built

environment; however, one clear issue is the

lack of understanding of the seismic behavior

of timber structures. The level of confidence

associated with the seismic analysis and

design of timber structures is much lower than

for reinforced concrete and steel structures. As

the result, timber structures are very

vulnerable during big earthquakes. In a

project, which is called "A Research ad

Development Project on Hybrid Timber

Building Structures" the main objective to

improve the structural performance is

summarized in the following three items.

1) Develop high performance hybrid timber

members consisting of timber and other

materials

1 Senior Research Engineer, Building Research Institute, Tsukuba-shi, Ibaraki-ken 305-0802

2 Group Leader, Building Research Institute

3 Head, National Institute for Land and Infrastructure Management

83
2) Develop high performance joints to connect

timber members ,and between timber and

other materials

3) Develop hybrid timber structures

incorporating timber with other types of

structures such as reinforced concrete or steel

frame.

In this paper, the current activities are

summarized.

On the other hand, it is very important to

improve seismic performance for existing

inferior and low-rise wooden buildings in

Japan. The project for their seismic

rehabilitation has started since 2002. The

research review and future plan also describes

in this paper.

2. HYBRID TIMBER STRUCTURAL

SYSTEM

The effective utilization of timber is urged

from the point of preservation of natural

resources and global environment also in

Japan. And the effective mixture of timber and

other materials is expected to extend the

possibility of building structures because of

the possibility to realize high performance in

both structural safety and fire safety. The

Building Standard Law of Japan was revised

in June 1998 and the potential to accept largescale and high-rise timber buildings with its

provision of performance based code. The

objectives of this research and development

project are, to develop high-performance

hybrid timber members and hybrid timber

structural systems, to develop the performance

evaluation methods for structural and fire

safety of these buildings, and to develop the

design methods for typical hybrid timber

buildings.

This project started in 1999 as a five-year

national project. There are some preceding

researches or designs in this area, such as

timber beams reinforced by steel or fibers,

floor systems with timber and reinforced

concrete, structural systems with reinforced

concrete or steel core structure surrounded by

timber frames, and so on. These preceding

researches are, of course, very useful for our

starting of the projects. However, one of the

most important issues for the structural

performance of hybrid timber structure in

Japan is the dynamic behavior of the

structures against strong grand motions. There

are some research needs adding these

preceding researches from the point of seismic

performance of such kind of buildings.

2.1 Research Plan and Organization

2.1.1 Research Organization

Fig. 1 shows the original organization of

research program. The members of the

research committee are scholars and persons

of experience, structural engineers, and from

the Ministry. This committee adjusts and

unifies all research items of each subcommittee into themes of the project. The

sub-committees carry out concrete research

and development on hybrid timber buildings.

The working groups were established in order

to solve efficiently each item of research and

development. Until fiscal year 2000, the

research committee, two sub-committees, and

six working groups were organized and the

sub-committee on fireproof stared from 2001.

84
2.1.2 Research Plan

Table 1 shows the annual plan in the every

fiscal year. In the first fiscal year several

investigations have been carried out to

confirm the feasibility and social needs of

hybrid timber buildings and to clarify the final

target of this project. And in fiscal year 2000,

concrete research activities have been started.

2.1.3 Final Targets

The final targets in the sub-committee on the

member and joint are summarized as:

Development of high-performance and highreliable hybrid timber members and joints,

Development of general test and evaluation

procedures for hybrid timber members and

joints,

Development of some design methods for

typical hybrid timber members and joints.

The final targets in the sub-committee on the

structural systems are summarized as:

Development of the rational and efficient

hybrid timber structures,

Development of general performance

evaluation method for hybrid timber structural

systems,

Development of some structural design

methods for typical hybrid timber structural

systems.

2.2 Research Themes and Tentative Findings

2.2.1 Investigations

In fiscal year 1999, following investigations

were executed to confirm the social needs and

the feasibility of hybrid timber buildings.

Investigation of demand for timber-based

middle-rise buildings,

Investigation of existing composite timberbased members, joints and structures,

Investigation of patents on composite timberbased members, joints and structures,

Investigation of oversea and international

standards and codes,

Review of the background of provisions for

timber buildings in Japanese current

regulations.

2.2.2 Hybrid Members

In the WG on Members, hybrid timber

members, such as glulam reinforced by steel

or fiber, are classified and their performance

was roughly evaluated reviewing the

preceding researches. And the theoretical

equations to predict generally the strength and

equivalent stiffness were derived. Now a

series of static loading tests and creep tests on

several types of hybrid beams are prepared to

validate the theoretical equations. Table 2

shows the outline of the test plan.

For the evaluation of structural performance

and the design methods of these members,

probabilistic approach seems essential due to

the variability of properties in timber-based

materials.

2.2.3 Joints Between Members

Joints between timber and steel, between

timber and reinforced concrete and joints

between hybrid timbers are treated in the subcommittee. Additionally, timber joints using

some new materials, such as carbon fiber, are

also investigated. At first, classification of

85
these joints were made according to the

location in buildings, component materials,

the stress transferred and so on.

In hybrid timber buildings of dual structure

type, joints with glued-in rods seem effective

joints to connect timber and other materials.

And in hybrid timber buildings such as middle

rise office buildings, the stress level may be

higher and more effective timber joints will be

required. Carbon fiber sheets or other

materials can be used for the reinforcement to

prevent the rapture perpendicular to grain.

Some series of tests on some types of joints to

connect timber and reinforced concrete and

some joints reinforced by fibers, rods, steel

plates or timber-based materials are now

prepared for tests.

Although creep behavior of these joints is

important problem for the structural

performance evaluation and design, there are

not so many researches on this matter. Now

the prediction method of the creep

deformation and the test plans are discussed in

the WG on joints between members.

2.2.4 Joints Between Structures

Joint system between structures is essential

research theme for the development of dual

structure systems. The WG on joint between

structures first picked up the problems which

occur in the joints between structures, such as

deference of creep behavior, deference of

shrinkage, deference of structural behavior

against lateral load, problems due to

condensation on the surface of concrete or

steel, and so on. Next the stress types and

stress levels were investigated for some

particular structural systems, and the effective

joint details were discussed.

2.2.5 Floor Systems

The WG on floor systems deals with various

kinds of floor systems, which are acceptable

for hybrid timber structures, such as timber

beams with plywood, timber beams with

reinforced concrete slab, timber deck, and

timber deck with reinforced concrete slab. In

the WG these floor systems were classified

and the performance of structure, fireproof,

sound insulation, etc. were roughly discussed.

For structural performance of floor systems,

not only the strength and stiffness against

vertical load, but also the shear strength and

stiffness in plane are discussed considering

the structural function as diaphragm against

wind and earthquake load. This function is

important especially, for example, in the

structural design with reinforced concrete core

systems later mentioned. And the distribution

of design earthquake load for these buildings

is an important research theme. Some studies

using time history analysis considering shear

deformation of horizontal frames are executed

to obtain proper earthquake design load. There

are not so many preceding researches for the

shear stiffness and strength in plane of floor

systems. Fig. 3 shows the plan of shear

loading tests on some floor systems.

2.2.6 Dual Structure Systems

Dual structure systems are considered to be

feasible hybrid timber structures. Fig. 4 shows

some conceptual models. As a case study,

building design is now tried on three types of

buildings, which are reinforced concrete core

86
system, reinforced concrete shear wall and

timber frame system, and system of timber

frames surrounded by steel frames. This case

study is useful to figure out the image of

target hybrid timber structures, and to find out

the problems for structural performance

evaluation and structural design methods.

2.2.7 Timber Based Structure Systems

Timber-based systems includes two types, one

is the structure system using hybrid timber

members and joints and another is the

combined system with different types of

timber structures, such as wall system and

frame system. In the WG on timber-based

systems, these hybrid systems were classified

and the problems for structural performance

evaluation are drawn out and arranged.

3. SEISMIC REHABILITATION OF

EXISTING WOODEN BUILDINGS

The origin of this project is the collaboration

research program with CUREE-Caltech

woodframe project, which had been

conducted from 1998 to 2001. The objective

was to improve the unreliable performance of

wooden houses. The following items were

examined by integrating current static and

dynamic tests, and developing new tests if

necessary.

1) Understand the dynamic collapse

mechanisms of wooden houses, 1-1) Collect

system testing data for analyzing seismic

behavior 1-2) Develop three dimensional

seismic analysis model (software) for wooden

houses

2) Control the catastrophic collapse of

wooden houses, 2-1) Develop joint and shear

wall systems to avoid a catastrophic collapse

2-2) Evaluate rehabilitation technologies

3) Propose international harmonious testing

protocol and evaluation of testing results, 3-1)

Testing protocol and evaluation for joints 3-2)

Testing protocol and evaluation for shear

walls

4) Discuss international harmonious design

procedure for wooden house, 4-1)

Specification procedure 4-2)Performancebased design procedure

In 2002, the main target was revised in

accordance with society problem. The seismic

rehabilitation systems are required as soon as

possible to ensure life safety. The typical

seismic rehabilitation methods and their

seismic performance evaluation procedure

will be proposed within the fiscal year 2002.

4. CONCLUSIONS

The project on hybrid timber structures started

in 1999 as a five-year national project. In

fiscal year 1999, some investigations were

executed to confirm the demand and

feasibility of middle-rise timber-based

buildings. In fiscal year 2000, concrete

research activities started in six working

groups. The main research themes are, general

equation for strength and stiffness of hybrid

members and joints, prediction of creep

behavior of hybrid members and joints, test

and evaluation methods for structural

performance of members and joints, structural

behavior of floor systems against both vertical

and lateral load, development of effective

hybrid structural systems, and enumeration

and solution of some problems in joints

87
between structures. The final targets of this

project are to establish general evaluation

methods for structural and fireproof

performance of hybrid timber buildings and to

make some design methods for typical hybrid

timber buildings. It will enable to construct

middle-rise and large-scale hybrid timber

buildings in general and also to stem global

warming by minimizing emission of carbon

dioxide. It is also important to improve the

seismic performance of existing inferior and

row-rise wooden structures. Their

rehabilitation scheme began from the fiscal

year 2002.

5. Acknowledgements

Authors would like to thank all the members

of research committees (Chairperson: Isao

Sakamoto of University of Tokyo), subcommittees and working groups under the

research and development project on hybrid

timber building structures.

BRI Project Team

MLIT

BRI

Cooperative

Research

Public Service Corporations

BCJ, BCS, JSCA,

SC on Members & Joints

WG on Members

WG on Joints between

Research Committee Cooperative Research i

II

L SC on Structural Systems

Chair: Prof. 1. Sakamoto

n\ 0 SC on Fireproof i

J

WG on Joints between Structures WG on Timber

based Systems

WG on Floor Systems

WG on Dual

Structure Systems

WG

WG

Collaboration

FFPRI NIED Universities

BRI: Building Research Institute, MLIT: Ministry of Land, Infrastructure and Transport, BCJ: The Building Center of Japan,

BCS: Building Contractors Society, JSCA: Japan Structural Consultants Association, HOWTEC: The Foundation of Japan

Housing and Wood Technology Center, FFPRI: Forestry and Forest Products Research Institute, NDED: National Research

Institute for Earth science and Disaster Prevention

Fig. IResearch Organization

88
Table 1 Research Plan

ITEMS F.Y. 1999 2000 2001 2002 2003

Development of members & joints

1) Investigation of current status

2) Research on mechanical behavior

- Creep Tests

- Tests on mechanical behavior

- Evaluation of mechanical performance

3) Development and analysis

- Proposal of basic concept

- Development of members & joints

- Tests & analysis

4) Research on fireproof performance

- Clarification of performance requirement

- Performance evaluation of members & joints

5) Design methods & performance evaluation

methods

- Establishment of basic concept

- Proposal

Development of hybrid timber structural systems

1) Investigation of current status

2) Decision of target structural systems

- Enumeration of hybrid structural systems

- Case study for enumerated systems

- Decision of target structural systems

3) Development and analysis

- Basic analysis for proposed systems

- Basic tests for proposed systems

- Full scale tests for proposed systems

4) Research on fireproof performance

- Clarification of performance requirement

- Performance evaluation of proposed systems

5) Design methods & performance evaluation

methods

- Establishment of basic concept

- Proposal

Table 2 Test Plan ofHybrid Timber Members

Type of Member Section of

Glulam

(mm)

Section of

reinforcement

(mm)

Connector or Adhesive

(mm)

Span of 3-point

bending

(mm)

Glulam beam

reinforced by steel

in bith sides

100x200 6x65 2-LS*(d=9, l=75)@70 4,000 and 2,000

LS (d=9, l=75)@70 4,000

LS (d=9, l=75)@140 4,000 and 2,000

Glulam beam

reinforced by steel in tension sides

100x200 6x65 2-LS (d=9, l=75)@70

LS (d=9, l=75)@70

LS (d=9, l=75)@140

4,000

Glulam Beam

reinforced by carbon

fiber sheet in bith

sides

D

100x200 1.4x100 Resorcinol 4,000 and 2,000

2.8x100 4,000

4.2x100 4,000 and 2,000

Glulam Beam

reinforced by carbon

fiber sheet in tension

side

D

100x200 1.4x100

2.8x100

4.2x100

Resorcinol 4,000

*LS: Lugscrew

89
3* DC

Ha

K3,640

CP"

A: Glulam Beam and Plywood

n—n—n—tr i]

B: Wood Deck

C: Glulam and Reinforced Concrete

ff u u g4

D: Wood Deck and Reinforced Concrete <p

[jimilimiiniiii Miliiiiiiiiiiiiiiiitiiiiii liiiiiHMhBiiatW^] y

Fig. 3Test plan on floor systemsfor shear stiffness and strength in plane

Reinforced concrete

Fig. 4 Conceptual Models ofDual Structural Systems

90
Modeling Regional Economic Resiliency to Earthquakes:

A Computable General Equilibrium Analysis of Lifeline Disruptions

by

Adam Rose and Shu-Yi Liao1

ABSTRACT

Business interruption losses from natural

hazards and terrorist attacks can be as costly as property damage. They can occur even when no

physical damage takes place, as when a business

is cut off from one of its utility lifelines.

Moreover, such disruptions set off a chain

reaction of further production cutbacks among

successive rounds of customers and suppliers

spreading through the entire regional economy.

This paper refines an advanced modeling

technique called computable general

equilibrium (CGE) analysis to make it suitable

for the assessment of the regional economic

impacts of lifeline disruptions in the aftermath

of a disaster. CGE is the state of the art of

regional economic impact analysis but has

several weaknesses. We develop a methodology

for recalibrating key model parameters in light

of empirical data on responses to lifeline

disruptions that reflect individual business and

regional resiliency. The methodology is applied

to analyzing the economic impacts of a

disruption to the Portland, Oregon water supply

system in the aftermath of an earthquake.

Comparison is made between impacts of a

business-as-usual scenario and one that reflects

the replacement of cast-iron pipe by advanced

materials.

KEYWORDS: economic impacts of disasters,

business interruption losses, resiliency,

computable general equilibrium analysis

1.0 INTRODUCTION

Recent studies indicate that utility lifeline

supply disruptions can have significant impacts

on regional economic activity in the aftermath

of an earthquake, other natural disaster, or

terrorist attack (see, e.g., Chang et al., 2000).

Even businesses that incur no physical damage

are likely to have to curtail their production if

they are cut off from their electricity, natural

gas, water, or communication links. Moreover,

such disruptions will set off a chain reaction of

further production cutbacks among successive

rounds of customers and suppliers spreading

through the entire regional economy. Surveys

following the Loma Prieta and Northridge

earthquakes, Hurricane Andrew, and the 1993

Midwest floods indicated that business

interruption losses stemming directly or

indirectly from lifeline failures rivaled property

damage in dollar terms (see Webb et al., 2000).

Research on this subject has until now been

dominated by the application of input-output

(I-O) models, which are linear, rigid and

characterized by simplistic responses; and by

1 The authors are, respectively, Professor and Head, Department of Energy, Environmental, and Mineral Economics,

The Pennsylvania State University, University Park, PA 16802; and Economist, Demand Analysis Office, California

Energy Commission, Sacramento, CA 95814. The research in this paper was supported by a grant from the National

Science Foundation (Grant No. 9802151), and by several grants from the NSF-sponsored Multidisciplinary Center

for Earthquake Engineering Research. The authors acknowledge the contribution of Chunsheng Shang, Gauri Guha,

Ram Ranjan, and Debo Oladosu for research related to the paper. We are also grateful to Stephanie Chang for

providing us with background information on the Portland Bureau of Water Works network and sharing with us

results of her analysis study of network vulnerability, direct economic impacts of earthquake disruptions, and

maintenance and mitigation costs. Helpful comments on an earlier draft were provided by Ann Fisher. The views

expressed in this paper of those of the authors and do not necessarily represent the positions of any sponsoring

organization or organizations to which the authors are affiliated. The authors are solely responsible for any errors and omissions.

91
mathematical programming models, which omit

any behavioral responses to prices and markets

(see, e.g., Cole, 1995; Rose et al., 1997).

Computable general equilibrium (CGE) analysis

would appear to be the superior alternative for

the next generation of regional economic impact

methodologies. It is comprehensive, non-linear,

based on individual behavior in response to

market prices, and explicitly incorporates a broad range of constraints (see, e.g., Rose, 2002;

Rose and Guha, 2002).

One of the few advantages of 1-0 over CGE

modeling is the clear distinction between direct

and indirect impacts in the former.

Interdependence is even stronger in the CGE

framework because of its price-quantity

interconnectedness. Thus, CGE results are

typically presented in terms of total impacts

only and are difficult to decompose. Another

shortcoming of CGE models is the lack of data

for estimating the many additional key

parameters they require, especially elasticities of

substitution.

The methodology in this paper provides an

alternative to the usual non-survey "adaptation"

approach to CGE model parameter estimation.

It utilizes primary data on direct economic

losses and a sophisticated computational

algorithm to recalibrate sectoral production

functions in response to a water lifeline service

disruption in the aftermath of a major

earthquake. It also measures how resiliency

improves when mitigation is implemented in the

form of replacing old cast-iron pipe with the

latest generation of durable materials. Our

modeling advances are inspired by the extensive

empirical work on business interruption in the

aftermath of earthquakes by Tierney (1997),

Chang (2001) and others. Our parameter

adjustments are linked to specific real world

examples of business resiliency (e.g.,

conservation, use of back-up supplies and

equipment, increased substitutability). The

approach also enables us to distinguish between

direct and indirect losses in a CGE context.

This refinement is important because, while

each of these adjustments may be capable of

reducing the economic shock to the same

i minimum direct reduction in output, each has

different implications for indirect impacts (see,

e.g., Rose and Lin, 1995).

2.0 CGE MODEL ADVANTAGES

Computable General Equilibrium (CGE)

analysis is the state-of-the-art in regional

economic modeling, especially for impact and

policy analysis. It is defined as a multi-market

simulation model based on the simultaneous

optimizing behavior of individual consumers

and firms, subject to economic account balances

and resource constraints (see, e.g., Shoven and

Whalley, 1992). The CGE formulation

incorporates many of the best features of other

popular model forms, but without many of their

limitations (Rose, 1995).

The basic CGE model represents an excellent

framework for analyzing natural hazard impacts

and policy responses (Boisvert, 1992;

Brookshire and McKee, 1992; Rose and Guha,

2002). CGE models can be finely

disagggregated to better distinguish the various

degrees of vulnerability to hazards across

sectors. The production functions are inclusive

of all inputs, not just primary factors as in the case of many other economic models, which

facilitates identification of materials shortages.

At the same time, CGE models allow for the

possibility of input substitution, which mimics

real world responses beyond the very short run

in minimizing hazard impacts. They also allow

for the substitution of imported goods for

regionally produced goods. CGE models are non-linear in form, thereby more closely

reflecting real world conditions, such as economies of scale and non-linear damage

functions. CGE models are more capable of

analyzing disjoint change than are model forms

based on time series data and which therefore

simply extrapolate the past. They can also more

readily accommodate engineering data or data

based on informed judgment.

Another set of CGE model advantages pertains to

the important role of prices and markets. Related

to this is explicit consideration of behavioral

response, and not just simple optimization but also

92
instances of bounded rationality, applicable to

both mitigation and recovery behavior.

Finally, CGE models are superior to some other

alternatives in modeling the role of lifelines and

infrastructure, which include societal necessities

such as electricity, water delivery,

transportation, and communication. Many of

these services are especially vulnerable because

they are provided in linear links so that if one is disrupted, so are all downstream extensions. A

CGE model can place a valuation on these

services, even for public sector outputs that are

typically unpriced. This is more than an academic exercise, because "shadow values"

might serve as temporary prices to ration these

services through the market rather than by

administrative decree (Rose and Benavides,

1999).

3.0 RESPONSES TO HAZARDS IN A CGE

CONTEXT

The production side of the CGE model used in

this paper is composed of a multi-layered, or

multi-tiered, constant elasticity of substitution

(CES) production function for each sector. The

CES has several advantages over more basic

forms such as the Leontief (linear) or CobbDouglas (simple multiplicative) functions (see,

e.g., Perroni and Rutherford, 1993). It can

incorporate a range of input substitution

possibilities (not just the zero and unitary values

of the aforementioned functions). The multiple

tiers allow for the use of different substitution

elasticities for different pairs of inputs (the

elasticity is constant for a given tier, but

elasticities can vary across tiers). The

production function is normally applied to aggregate categories of major inputs of capital,

labor, energy, and materials, with subaggregates possible for each (e.g., the energy

aggregate is often decomposed by fuel type

electricity, oil, gas, and coal). Water is usually

omitted or incorporated as one of the materials

(intermediate goods producing) sectors. We

explicitly separate water as a major aggregate in

the top tier of the production function so that we

can analyze the impacts of a water service

disruption. Again, our methodology provides a

way of improving the empirical estimates of key

elasticities.

3.1 CES Production Function

Our constant elasticity of substitution (CES)

production function has the following nested

form for five aggregate inputs capital, labor,

energy, materials, and water:

Y = Afa Aw W~Px +J3l KLEM~p> y

Px (1)

KLEM=A2 (a2 M~Pl +p2 KEL'Pl ^

Pl (2)

KEL=A3 ^3 L'

p3 +/?3 KE'P3 Y

P3

(3)

KE=A4 (a, K~p * +/34 E~P

< J

p* (4)

where:

Aj is the factor-neutral technology parameter,

Ai>0

Aw is the water-specific technology parameter

ahPi are the factor shares, 0<a„ /?, <1

<Ji is constant elasticity of substitution,

1 0/=:

l+Pi

Y is output

K,L,E,M, W are individual capital, labor, energy,

material and water aggregates

KLEM is the capital, labor, energy, and

material combination

KEL is the capital, energy and labor

combination

KE is the capital and energy combination

The fixed coefficient production function of an

1-0 model would yield an upper-bound estimate

of direct output losses from water input

disruption, where the percentage loss of the

former would be equal to the percentage loss for

the latter. All other types of production

functions would yield percentage output losses

lower than the percentage decrease in water

availability because of substitution possibilities.

93
We define individual business resiliency as the

difference between the fixed coefficient

(proportional) result and the flexible input

(disproportional) result, and which is attributable both to the various response

mechanisms related to water services (1st Tier)

and inherent in the overall production function

with respect to other inputs (Tiers 2-4).

CGE models used for hazard analysis are likely

to yield estimates of business disruptions for

some if not all sectors of an economy that differ

significantly from the direct loss estimates

provided by empirical studies. This is because

production function parameters are not typically

based on solid data, or, even where they are, the

data stem from ordinary operating experience

rather than from emergency situations. Hence,

it is necessary to explicitly incorporate the

resiliency responses below into the analysis.

This is accomplished here by altering the

parameters, and, in one case, the variables in the

sectoral production functions of the CGE model.

3.2 Production Responses to Natural Hazards

Below we summarize types of response to natural hazards, linked to the production

function tier and parameters to which each

relates and to the recovery/reconstruction stage

(time period) to which each is applicable (see

Rose and Liao, 2002, for the mathematical

derivation of the major production function

responses). These responses include:

1 . Conservation of Water. This response can be

implemented immediately and continued

through the long-run, i.e., be incorporated into

the production process on a permanent basis.

One of the silver linings of disasters is that they

force businesses to reconsider their use of

resources, and often not just at the margin for a

single input but also holistically (as noted in

item 7 below). The parameter changes for this

response in the case of water pertain to the

technology trend variable in the first tier of the

production function specified above. More

generally, in each tier of the production

function, the productivity term, Ah is specified

as covering all inputs, i.e., factor neutral.

Adjustment of the productivity term for an

individual factor, such as the A\ W term in

equation (la), biases the productivity

improvement in the direction of that factor.

2. Conservation of Other Inputs. This

is analogous to water conservation and can be

applied to any of the tiers. However, it can

often take on more permanence than water

conservation, which is a dire necessity in many

cases, and it is listed in Table 1 as constant over

the applicable period rather than decreasing.

Examples would include a reduction in number

of trucks or maintenance personnel. One other

adjustment option can be thought of as a sub- case—an increase in the use of non-water

inventories, though only through the very short

run.

2

3. Increased Substitutability of Other Inputs for

Water System Deliveries. This response would

be exemplified primarily by purchasing water

from other sources (by the bottle or truckload), or by moving to another location where less

water is needed.

4. Back-up Supplies . This response is often

implemented in the immediate aftermath of an

earthquake in the short-run. It includes

adjustments that incur costs, such as the digging

of wells, and rather costless measures, such as collecting rainfall or using riverine water. The

costless alternatives can be modeled in a manner

similar to conservation and the cost-incurring ones similar to substitution. The use of water

inventories (stored water) is best addressed as discussed above. As with the inventory item

discussed above, there is some flexibility in how

costs are considered temporally.

5. Water Importance. This response requires

more explanation because of the widespread use

2 Actually, an increase in inventory use may be more

like response number 5 below because material is

actually used and not saved; however, it is listed here

because the purchase has been made in an earlier time

period (and will likely be replenished in later one).

94
of the term "importance" in its broadest sense in

the earthquake research literature. Sometimes,

it has been used to encompass all of the

responses noted in Table 1. In ATC-25 (1991),

utility lifeline importance was quantified as the

percentage change in a sector's output that

would result from a one percent change in input

availability. If water were used everywhere in

the production process and no resiliency measures were possible, a one percent decrease

in water would lead to a one percent decrease in

output, or an importance factor of 1 .0 (the same

as the 1-0 fixed coefficient production function).

The existence of various responses lowers the

importance factor, which had a value as low as

.30 for the Transportation and Warehouse sector

in ATC-25. Here we go to the opposite extreme

in the use of the term as the percentage of

production activities in a given sector that do

not require water to operate. Thus, it refers to

the inherent resiliency of a production process

in the absence of any explicit adjustment. The

presence of this factor is labeled as constant in

Table 1; any increase in it over time would

mean further technology adjustment and would

come under the headings of responses 1 or 7.

3

6. Time-of-Day Usage. This is a passive

response that pertains to hours during which the

business is closed, and hence where loss of

water has no effect on output (see, e.g., Rose

and Lim, 1997, for an example of how this

adjustment greatly reduced loss estimates from

electricity disruptions in the aftermath of the

Northridge Earthquake). It is listed here for the

sake of comprehensiveness.4

3 Note that ATC-25 assumed that the first 5 percent of

lifeline service disruption would not result in any loss

of output by customers due to various resiliency

measures, such as the others noted in Table 1 . Hence,

there is some ambiguity about the extent to which the

ATC-25 importance factor is intended to be limited

just to separable production activities as in the

definition in this paper.

4 In a more extensive model involving dispatch

decisions in which water or electricity might be

rationed by time of day to shave peak production, this adjustment comes into play for firms that cannot

switch production to off-hours.

7. Change in Technology. This refers to long run (permanent) changes in the overall process,

such as replacing open systems, which do not

recycle water, with closed systems. It may

require the reformulation of the entire

production function.

4.0 PORTLAND ECONOMY AND WATER

SYSTEM

4.1. The Portland Water System

The Portland Bureau of Water Works (PBWW)

is a rate-financed, City-owned utility that serves 840,000 people in portions of the Portland

Metro Area (including businesses responsible

for 98% and 72% of sales in Multnomah County

and Washington County, respectively). In 1999,

PBWW water sales amounted to 39 billion

gallons. The largest customers are major

manufacturing companies, the Portland City

Bureau of Parks and Recreation, and several

hospitals.

The PBWW transmission and distribution is

comprised of nearly 2000 kilometers of

pipelines, 29 pump stations, and 69 major

storage tanks. Construction of the PBWW dates

back to 1894. About 70% of the system still consists of cast iron pipes, even though the

agency began installing ductile iron in the

1960s. Additional information on the PBWW,

its maintenance and earthquake mitigation costs,

and its earthquake vulnerability can be found in

Chang (2001).

4.2 Portland CGE Model

We constructed a CGE model of the portion of

Portland Metropolitan Area economy that

overlaps with the Portland Bureau of Water

Works (PBWW) Service Area. The main data

upon which the empirical model is based are the

1998 IMPLAN Social Accounting Matrix

(SAM) and Input-Output Table for Multnomah

County and Washington County (MIG, 2000).

It is divided into several partitions that reveal

95
the structure of the regional economy, including

the industry, commodity, factor income,

household, government, capital, and trade

accounts.

The SAM industry accounts contain 20 sectors,

with the Water & Sanitary Services separated

from other utility services in order to pinpoint

economic impacts of water supply disruptions in

the aftermath of an earthquake. The Total Gross

Output of the Portland Metro economy in 1998

is $71.2 billion, including $42.1 billion in

interindustry transactions and $29.1 billion of

total value-added. The total domestic

commodity supply and exports of the Portland

Metro Area in 1998 are $43.3 billion and $27.9

billion, respectively, implying the region is moderately self-sufficient. This is further

evidenced by the trade accounts. The net

domestic trading surplus is about $3.4 billion

and the net foreign trading deficit is about $5.2

billion. Major features of the Portland CGE

model are described in Rose and Liao (2002).

4.3 Model Solution Algorithms

Previous research on the application of CGE

models to lifeline disruptions (see, e.g., Rose

and Guha, 2001) has made use of the basic

nonlinear programming features of the General

Algebraic Modeling System (GAMS) software

(Brooke et al., 1988). However, GAMS was not

able to incorporate the range of substitution

elasticity values required. We constructed the

Portland CGE model using MPSGE, a subsystem of GAMS that facilitates the

formulation and analysis of applied general

equilibrium models through complementarity

programming (Rutherford, 1995), and which

allows for a broader range of parameter values

and the inclusion of a broader range of

constraints (e.g., sectorally differentiated supply

availabilities of lifeline services). Basic

elasticity of substitution values in the model

were based on a careful synthesis of the

literature (see Rose and Guha, 2002). Because

we are modeling a very short-run response we

further modified all input and trade substitution

elasticities (except input elasticities pertaining

to water as discussed below) so that they were

10% of their initial values.

The recalibration of the elasticity of substitution

(pi) between water (W) and the capital, energy,

labor, and material aggregate {KELM) could

only be undertaken with a numerical solution, in

this case the numerical bisection method, which

is a converging root search routine.

5 Given

sectoral water availability and the corresponding

output reduction rate, the recalibration involves

three iterative steps: First, find a lower bound

of oj . To obtain an initial unadjusted output

reduction rate higher than Chang's estimate for

all sectors, 0.05 was used as the lower bound for

all sectors. Second, find an upper bound of oj

.

Our initial guess of the upper bound of o\ was

0. 15. which yielded a corresponding output

reduction rate lower than Chang's estimate for

all sectors. Third, calculate the arithmetic

average of the lower and upper bounds of o\

and its corresponding output reduction rate. If

the output reduction rate is higher (lower) than

the direct output loss estimate, then the

arithmetic average can be used as the new lower

(upper) bound of oj for the next iteration. The

iterative procedure is continued until the

deviation between the calculated output

reduction rate and the empirical direct loss

estimate is less than 0.01%. In our model, the

routine converged to the recalibrated 05 in less

than 10 iterations for each sector.

The computational procedure we have

developed to improve model accuracy also

generates an additional dividend of enabling us

to decompose loss estimates. Our estimation of

indirect losses involves the following procedure:

1. Extract the sectoral production functions

from the CGE model and adjust parameters

and variables in them one at a time to match

empirical direct loss estimates.

2. Reinsert the recalibrated sectoral production

functions into the CGE model, reduce input

Other parameters, such as A w, can be solved

analytically (see Rose and Liao, 2002).

96
supply to a level consistent with empirical

estimates, and compute total regional losses.

3. Subtract direct losses from total losses to

determine indirect losses.

5.0. SIMULATING THE RESPONSE TO

NATURAL HAZARDS

5.1 Earthquake Parameters

The Portland Area is characterized by moderate

seismic activity stemming from the ocean floor

Cascadian Subduction Zone and a series of

shallow crustal faults. Two damaging

earthquakes have taken place in the past 40

years measuring M5.5 and M5.6. However,

large subduction earthquakes as great as M9.0

have taken place as recently as 1 700 (see Wong

et al., 2000).

Although PBWW service has not been disrupted

by earthquakes, a recent study by EQE (1999)

found that operators of the system consider it vulnerable. The study also identified mitigation measures that might help meet the System's

safety performance standards desired by

stakeholders.

Chang (2001) simulated the effects of three

alternative mitigation measures (no action, castiron pipe replacement, and tank/pump upgrade).

The analysis was undertaken in the context of a

life-cycle cost model that factored in not only

the cost of mitigation over time and its ability to

reduce system vulnerability through the year

2050, but also the savings of ordinary

maintenance costs. In the analysis below, we

will confine our attention to the Year 2000.

Also, we will focus on the direct and indirect

effects of system disruptions before and after

mitigation (through cast-iron pipe replacement

by the latest generation of advanced materials)

and will leave to another paper our analysis of

the regional impacts of implementing the

mitigation measures.

6

Chang (2001) estimates the cost of tank pump

upgrades as only $2.1 million, but that of pipe

replacement as $380 million.

5.2 Empirical Measurement of Resiliency

Chang (2001) performed simulations for

alternative combinations of earthquake types,

calendar years, and mitigation options, using

several sophisticated geological and engineering

models. Each case was subject to 100 Monte

Carlo simulations. These simulations were used

to estimate direct losses in sectoral output,

factoring in resiliency. Based on the work by

ATC (1991) and Tierney (1997), resiliency is defined by Chang as "the remaining percentage

of output that an industry can still produce in the

event of total water outage." Sectoral resiliency

measures range from a low of 21 percent for

Health Services to a high of 49 percent for

Transportation and for Communications and

Utilities. Note that the definition of resiliency

we provided above is a generalization of

Chang's definition to cases where the water

outage is not a total one. Note also that the

ATC definition assumes a linear relationship,

but that non-linear relationships are likely to be

more realistic. Our analysis below can be used

to estimate non-linear relationships between

water service disruptions and output reduction

and hence represents a non-linear measure of

resiliency.

6.0 WATER DISRUPTION SIMULATION

RESULTS

Simulations were conducted of the regional

economic impact of an earthquake-induced

water supply disruption in the Portland Metro

Area. The analysis is based on the simulation of

the systems analysis of the Portland water utility

system and the direct loss estimation

simulations of Chang (2001) described above.

Although Chang's engineering vulnerability and

direct loss simulations involve many scenarios

relating to alternative earthquake magnitudes,

outage durations, and resiliency responses, this

paper focuses on a subset of scenarios

characterized by:

97
1. One earthquake type (Bolton crustal fault)

of magnitude 6. 1

.

2. Impacts in the Year 2000.

3. Scenarios for Business as Usual (No

Mitigation) and Cast-iron Pipe

Replacement.

4. Outages of varying lengths from 3 to 9

weeks.

5. Resiliency responses involving only the

substitution of other inputs for water

services.

We focused on the first characteristic because it represented the "most likely" case and

characteristics 2-4 to keep the number of

simulations manageable. We confined ourselves

to the characteristic 5 because we have not yet

fully developed the methodology to implement

changes in other parameters.

Note, one other important dimension of our

simulations, which relates to pricing of water

delivery. Ordinarily, CGE simulations allow

prices to fluctuate freely in response to changing

supply and demand conditions. However, two

features of this situation warrant simulations

with fixed water prices. The first is the fact that

businesses often resist raising their prices in the

aftermath of a natural disaster for reasons of

altruistic community concern and to avoid the

image of price gouging. Second, the PBWW is not a typical enterprise with fluctuating prices

but rather one in which rates are adjusted only

periodically in the context of open public

hearings.

6.1. No Pre-Event Mitigation; Post-Event Water

Substitution

The results of our simulations for the Business

as Usual Scenario (no mitigation) are presented

in Table 1 . Note that the duration of this outage

is projected to be 4 weeks, but the table

summarizes the situation for the maximum

disruption, which takes place during the first week. Results for the total outage period are summarized more briefly below.

The sector labels on the left-hand side of Table

1 refer to the economic producing units of the

Portland CGE model, and direct water

disruption for each sector is presented in column

2. In 2000, for example, unmitigated sectoral

water disruptions sum to a 50.5 percent. As

discussed further below, negative indirect

effects on sectors such as Construction are so

great as to reduce water demand significantly

below the post-earthquake availability levels.

Baseline output is presented in Column 3 and

reflects the relative prominence of sectors in

Portland Metro economy; it also serves as a

reference point for our impact simulations.

Note that the Water sector (11) gross output

represents only 0.44 percent of the regional

economy.

Column 4 of Table 1 presents the direct output

losses equaling 49.1 percent that are estimated

in our model before any resiliency adjustment.

Chang's estimates of direct output losses

amounting to only 33.7 percent, which

incorporate the extent of resiliency, are presented in Column 5.

7 Our direct loss

estimates are based on input substitution

elasticities of 0.05 presented in the next to the

last column in Table 1 . Note that our direct loss

estimates exceed those of Chang in every sector

because ours omit all resiliency options except

normal input substitution (and at very low

levels). The final column shows the elasticities

necessary to incorporate resiliency measures for our model results to be consistent with the

Chang estimates (these range from a low

increase of 36 percent in Sector 18 to a high

increase of 130 percent in Sector 1 1 itself).

Our estimates of the indirect and total regional

economic impacts of the water lifeline

disruption are presented in Columns 7 and 8. Overall, they yield a 10.7 percent indirect

reduction in regional gross output and a 44.4

percent total reduction in regional gross output

One measure of resiliency would be the extent to which the actual direct output reduction deviated

from the likely (fixed-coefficient) maximum, which is equivalent to the percentage water input disruption.

The measure would be 33.3 percent in this scenario

[(50.5-33.7) + 50.5].

98
for the week. The former represents $586

million and the latter $2,431 million. Chang

(2001) assumes that restoration takes place in a

straight-line manner, so the total loss in

economic output for the Region is estimated to be $4,862 million. Moreover, for an actual

earthquake, damage to other utilities and

buildings would likely lead to losses much

greater than this, though care needs to be taken

to avoid double-counting losses to the same

business from several sources (see Chang et al.,

1996).

Some interesting aspects of general equilibrium

(indirect) losses for the first case are indicated

by Table 1. First, they are only about 32 percent

the size of direct losses. In the context of an

input-output model, this would be a multiplier

of only about 1.32. The Portland Metro

economy-wide output multiplier is significantly

larger than this, but the CGE model incorporates

many other factors that mute the uni-directional

and linear nature of the pure interdependence

effect of the 1-0 model. For example, it is able

to capture price changes for intermediate goods

from cost and demand pressures, various

substitutions aside from those relating to water,

and various income, substitution and spending

considerations on the consumer side. The sector

suffering the largest indirect decline is Construction (a leading indicator of economic

activity); however, if post-earthquake recovery

and reconstruction were to be factored in, the

Construction decline would be offset

significantly. In addition, several sectors are characterized by positive or minimally negative

indirect effects, most notably basic necessities,

such as Food Processing and Health Services.

6.2 Pre-Event Pipe Replacement and Post-Event

Water Substitution

The results of the scenario of an M6.1 crustal

fault earthquake but with cast-iron pipe

replacement are presented in Table 2. We

realize this scenario is somewhat unrealistic

because it accelerates the replacement of castiron pipes at an unrealistic pace but believe this

provides a useful illustration of the potential

advantages of mitigation. Note also that the

results represent a lower bound of the gains in

this type of mitigation in future years, since pipe

replacement would be even more valuable as the

current system deteriorates, thereby incurring

increasingly higher ordinary maintenance cost

and likely greater earthquake vulnerability.

In this second Scenario, the direct water outage

is reduced from 50.5 percent to 31.0 percent

(see column 2 of Table 2). Our initial estimates

of direct output losses are 30.7 percent,

compared to Chang's empirical estimates of

21.3 percent.

8 The substitution elasticity

adjustments needed for the model to replicate

the Chang direct loss estimates are presented in

the final column of Table 2. They are lower

than the corresponding water input changes in

each sector because the direct output losses are projected to be lower in each.

Interestingly, our estimate of general

equilibrium ("indirect") losses in Scenario 2 is 11.2 percent, which is a 52.6 percent greater

than direct losses. Moreover, not only is the

percentage increase over direct losses higher in

Scenario 2 than in Scenario 1 but the absolute

level is as well. The latter is quite surprising at

first glance. It would seem to be an

impossibility, for example, in the context of an

1-0 model (where multiplier values are the same

at all scales). However, our CGE model is nonlinear. Secondly, we should not forget that we

have changed parameters (with respect to water

substitution), so even in an 1-0 context

multipliers would differ (though only slightly

given the small size of our parameter changes,

which would correspond to the coefficient

The resiliency index defined in endnote 7 is 31.3

percent in this case. Resiliency would seem to decrease from the 33.3 percent of Scenario 1, but this

is likely due to the fact that resiliency opportunities

decrease as the size of the disruption decreases (and

vice versa, though some threshold may exist for

especially large disruptions that overwhelm the

resiliency capabilities). Note also that mitigation

effectiveness, with respect to the difference in direct

water losses between the two scenarios, could be

measured by a similar index and would equal 38.6

percent [( 50.5-31) - 50.5].

99
changes that would be utilized in an 1-0 model).

The major explanation of the relatively higher

percentage of general equilibrium effects in

Scenario 2 is due to the difference in the

sectoral mix of direct water disruptions in

relation to Scenario 1. This changes relative

prices, and the model, despite its relatively low

substitution elasticities, is quite responsive. For

example, sectors 3, 5, and 7 all incur indirect

effects greater than direct ones, and these

sectors, as well as 6, 12, 16, and 20, incur

greater indirect losses in Scenario 2 than in

Scenario 1, despite lower direct losses in the

former.

Overall, Scenario 2 is estimated to incur a 32.5

percent loss in gross output in the Portland

Metro economy in the Year 2000 during the first

week of water service disruption. In dollar

terms, this translates into $1,800 million. Chang

(2001) estimates the system can be restored to

full service within three weeks in this case, so, again assuming a linear restoration path, total

output loss is $2,700 million, for an overall

reduction from business as usual of 44.5

percent. We now decompose this improvement

into its various constituents. The greatest

contribution of cast-iron pipe mitigation is the

reduction in restoration (-24.4 of the -44.5

percent) time, followed closely by the decrease

in direct losses (-21 percent). Indirect losses

(0.8 percent) are beyond the control of

policymakers, and, although they undercut the

effectiveness of this mitigation alternative in

this case, their influence is minimal. Moreover,

in most cases they are likely to be reinforcing.

9

Note that we also simulated cases (not shown) in

which water prices were allowed to fluctuate in

response to changing conditions of supply and

demand. Interestingly, despite a 33 percent increase

in water prices in this variant of Scenario 2 the

overall results were similar, though indirect effect

losses did decrease from 1 1.2 percent to 10.5 percent, or lower than the 10.7 percent of the Business as Usual Scenario. The reason for this improvement is

that resources are allocated more efficiently when

prices are not fixed, though the price increase will be

especially burdensome to marginal businesses and

low income consumers. This points out one of the

important equity-efficiency tradeoffs in the mitigation

7.0. CONCLUSION

This paper has presented major methodological

advances in computable general equilibrium

modeling for application to estimating the

regional economic impacts of earthquakes and

other disasters. First, we have provided an

operational methodology for recalibrating model

parameters in light of empirical estimates of

production losses due to a lifeline supply

disruption. Second, the methodology enables

the analyst to decompose loss estimates into

direct and indirect components. Our application

to a disruption of water services of the Portland

Metro economy showed that indirect economic

losses can vary substantially according to the

overall level and sectoral mix of water

shortages.

Overall, our results appear to be reasonable for

the economy as a whole, for individual sectors,

and for individual impact stages (direct and

indirect). We have, however, modeled only one

resiliency measure, albeit one of the most

important ones—increased substitutability of

other inputs for water utility services. As

discussed earlier, different types of resiliency

responses would generate different types of

indirect impacts. However, only a full simulation will provide a definitive conclusion.

of natural hazards impacts. Mitigation is often just

associated with pre-event activities, but it also

pertains to post-earthquake responses, such as reducing restoration times or modifying pricing, both

of which can have the effect of reducing overall

losses or losses to various stakeholders (see also Rose

et al., 1997; Shinozuka et al., 1998).

100
8.0 REFERENCES

Applied Technology Council (ATC). 1991.

Seismic Vulnerability and Impact of Disruption

of Lifelines in the Conterminous United States,

Report ATC-25. Redwood City, CA: ATC.

Boisvert R. 1992. Indirect losses from a catastrophic earthquake and the local, regional,

and national interest, in Indirect Economic

Consequences of a Catastrophic Earthquake,

Washington, DC: FEMAO.

Brooke A., Kendrick D, Meeraus A. 1988.

GAMS: A User's Guide. South San Francisco,

CA: The Scientific Press.

Brookshire D, McKee M. 1992. Other indirect

costs and losses from earthquakes: issues and

estimation," in Indirect Economic Consequences

of a Catastrophic Earthquake, Washington, DC:

FEMA.

Chang S. 2001. Evaluating social benefits of

disaster mitigation, Department of Geography,

University of Washington, Seattle, WA.

Chang S, Seligson H, Eguchi R. 1996.

Estimation of the economic impact of multiple

lifeline disruption: Memphis Light, Gas, and

Water Case Study, Buffalo, NY: National

Center for Earthquake Engineering Research.

Chang S, Rose A, Shinozuka M, Svekla W,

Tierney K. 2000. Modeling earthquake impacts

on urban lifeline systems: advances in

integration, in Research Progress and

Accomplishments, 1999-2000, Buffalo, NY:

Multidisciplinary Center for Earthquake

Engineering Research.

Cochrane H, et al. 1997. Indirect economic

losses, in Development of Standardized

Earthquake Loss Estimation Methodology Vol.

H, Menlo Park, CA: RMS, Inc.

Cole S. 1995. Lifelines and livelihood: a social

accounting matrix approach to calamity

preparedness, Journal of Contingencies and

Crisis Management 3: 1-11.

French S. 1998. Spatial analysis techniques for

linking physical damage to economic functions

and Shinozuka M, Rose A, Eguchi R (eds.),

Engineering and Socioeconomic Impacts of

Earthquakes: An Analysis of Electricity Lifeline

Disruptions in the New Madrid Area, Buffalo,

NY: MCEER.

(MIG) Minnesota EVIPLAN Group. 2000.

Impact Analysis for Planning System

(IMPLAN), Stillwater, MN.

Perroni C, Rutherford T. 1993. Regular

flexibility of nested ces functions, European

Economic Review 39: 335-43.

Prywes M. 1986. A nested ces approach to capital-energy substitution, Energy Economics

8: 22-28.

Rose A. 1995. Input-output economics and

computable general equilibrium models,

Structural Change and Economic Dynamics 6:

295-304.

Rose A. 2002. Model validation in estimating

higher-order economic losses from natural

hazards, in Taylor C, VanMarcke E (eds.)

Acceptable Risk to Lifeline Systems from

Natural Hazard Threats, New York: American

Society of Civil Engineers.

Rose A, Benavides J. 1999. Optimal allocation

of electricity after major earthquakes: market

mechanisms versus rationing, in Lawrence K

(ed.) Advances in Mathematical Programming

and Financial Planning, Greenwich, CT: JAI

Press.

Rose A, Guha G. 2002. Computable general

equilibrium modeling of electric utility lifeline

losses from earthquakes, Forthcoming in Chang

S, Okuyama Y (eds.), Modeling the Spatial

Economic Impacts of Natural Hazards

Heidelberg: Springer

Rose A, Liao S. 2002. Modeling regional

economic resiliency to earthquakes: a computable general equilibrium analysis of

101
lifeline disruptions, Department of Energy,

Environmental, and Mineral Economics, The

Pennsylvania State University, University Park,

PA.

Rose A, Lin S. 1995. Regrets or no regrets

that is the question: is conservation a costless

C02 mitigation strategy? Energy Journal 16: 67-

87.

Rose A, Benavides J, Chang S, Szczesniak P,

Lim D. 1997. The regional economic impact of

an earthquake: direct and indirect effects of

electricity lifeline disruptions, Journal of

Regional Science 37: 437-58.

Rose A, Liao S, Ranjan R, Guha G. 2001.

Estimating economic losses from earthquakes: a computable general equilibrium analysis of the

effects of electric utility lifeline losses,

Department of Energy, Environmental, and

Mineral Economics, The Pennsylvania State

University, University Park, PA.

Rutherford T. 1995. Computable general

equilibrium modeling with MPSGE as a GAMS

subsystem: an overview of the modeling

framework and syntax, http://www.gams.com/

solbers/mpsge/svntax.htm.

Shinozuka M, Rose A, Eguchi R/ (eds.). 1998.

Engineering and Socioeconomic Impacts of

Earthquakes, Buffalo, NY: Multidisciplinary

Center for Earthquake Engineering Research.

Shoven J, Whalley J. 1992. Applying General

Equilibrium, New York: Cambridge University

Press.

Tierney K. 1997. Impacts of recent disasters on

businesses: the 1993 midwest floods and the

1994 northridge earthquake, in Jones B (ed.),

Economic Consequences of Earthquakes:

Preparing for the Unexpected, Buffalo, NY:

National Center for Earthquake Engineering

Research.

Webb G, Tierney K, Dahlhamer J. 2000.

Businesses and disasters: empirical patterns and

unanswered questions, Natural Hazards Review

1: 83-90.

Wong I, et al. 2000. Earthquake scenarios and

probabilistic ground shaking maps for the

Portland, Oregon, Metropolitan Area,

Interpretative Maps Series IMS- 16, State of

Oregon, Department of Geology and Mineral

Industries, Portland, OR.

102
c

1 i

OH

13

60 O

!3 .a *(J Q

o ^

.—

to c c

CQ

T3

c ~ .2 ° 5

B §• S ab

<u .a b

CQ

o

oo os r- CO <•<"> oo m in oo in m oo fN oo f- oo m roo oo oo Os 00 oo oo r- OO r- r- vo r-- q q q q q q q q q q q q q q q q q q q

© o o © © © ©' o o o o o o o o o o o o o

c_p wP —w ^

' f~> in IT) in in in m m in in in m m m in in m m m

q q q q q q q q q q q q q q q q q q q q

o o o o o o o o o o o o o o o o o o o o

0

s- xO

0

s- -o0

s- xP0s 0s 0s sP0

s- 0s- 0s- NP0s sPo^

- o

0s nP0s nP0s \p0

s- sP0s- sP0s sP0s- >o q o r-- C\ ON q q Is- in in "* OS VO fN m m

in 00 os (N in OS in oo vd rn r<S li-i fN in d> fN

f*1

i CO ini i mi in

i -J t*^i CNi i CO1 CO fn

i f*"li 1

-4-

XJ

sP

5

s- 0

s- xO0

s- 0s

- o

0s xO xP

0

s- xQ

0s vO0s 0s sP

0

s- sP

o^ nP

0s nP0s nPON nP0s- vP0s -O0s \00s sp0s oo in OO l> r- 00 in in m r-- ^* o tN O fn r-- vd © oo o «N fN o oo fN o

1 1

rxl

1 i v N1

1 1 I

1 1

1 1 1

0

s- os \=

cr- s°os N?0s v=0s- S=> 0s- 0s 0s sP0s os 0s- 0s os -°0s- -p0s s=0s- N=

0

s- N?0s ^p0s- sP0

s- m 00 VJ (N fN q q q q q VO VO >n O r~; fi v© ci v© VO in in in vd vd in fN vd rn m fN

i i m f>i i tN

i fN

i i fN

i mi i i • i 1 m mi

0s s=

ox

so 0s- 0s s=

0

s- N=0s- 0s ^?0s ox

sPoN 0

s- 0

s- v°

0

s- 0 s sP

0

s- o^ N?0s ox 6s- -p0s O t-; >o q in in Os fN oo Os Os Os fN q r~- Os

MO v© OS oo <N <N in vd vd >d vd OS CO vd od Os 00 o Os Os

"* m in * -* ^1- in

"* ^ S r-^ o fNi os K Os in vo vo o fN tN —' oo m —

in — vo

r-r--r--c-> — ooi^^orNo

r^i tN o in fN oo inosr— — fNOsmtN

os in m —i Os —

— — h -t O

—< *t vq os m tN - - o\ vo S ->

r--. in o oo oo

in os in os — r- m rn

, O ,0 .O sO yfl sO nP \0 -O -O .O vO \P sP vP \8 xp vfl xp q\ qX qX q\ q\ g\ qS q-x qN o Cs 0s- 0s 0s 0

s* Os 0 s Os Os Os ^Os00^oooin^ininos0sqosqfnos0sr<i OOCOOOOs^OOOOOOOOf'i^OOsOOOfN'—'OS ^tTj-voinininrj-TfTi-'^-mininTl-inininmin'^- i i i i i i i i i i i i i i i i i i i l

XPxPxPvPx=X?X?xPxOxPx°X?X°xOxOxOx°X=xOxO 0s- 0s- os os 0s- os 0s 0s- 0s Os 0s os 0s os os o> 0s- os os os OvomOsinmooOOOOOOOsOsOsfNOrss Osr<i

OO 00

Os O

O O

Os-^^r-oooooooo — ooo O — O fN —i Os

in in m in in

xt — oo Tj" oo fN

oo m rtN O r-- —i fN - a n - H d •* csi m — r- in in oo vo N oi m Is d o\

Os

oo

oo

0sO

oo

o

§ g

3 ^2

rS 6

Er W

•a o

'p OO a .s

a, ^

e s

S o

103
b

4—

»

'5

-252

c

u

<

3 s

a/

3

o

C3

£

o—

U*-» 3

fi^ 3

O

oH

u

XIc

U500 O

3 <L> «3 .3

§ £

S o o ^

.S 3

fi^ .2 ? S3 3

3

34

^5

co

a.3

CD .S 3

« .2

C/3 —

c3 r3

OQ £

o

oo

xi- VI CO o vo oo SO VO SO (SI o <T( NO NO so so NO SO OO NO SO so so so so SO SO SO O o O o o P o o o o o o o O o O o o O o

© o d d o d d d d d d d d d p d d d d d

o o o o o o © o o o o o o o o o o o o o io in u"N. in *^ *o v> V"l o o o o o o o o o o o o o o o o o o o o

o d d d d d d d d d d d d d d P p d d d

2s 2s Ss© ©^ 2? 2? 2? 2? 2?os 2s 2S 2? 2s 2s 2so 2s 2?o^ 2? OS

On no o> co CN 00 OO o ON CS) CO in so

C< r^i o\ VO oo d GO OS rsi

i 1 i t CO

1 I 1 CO1 1 i CO1 1 1 CO CO1

5? 2? s? 2? 2? 2? 2s 2s 2? 2? 2? 2? 2? 2? 2? 2s 2s 2?

csl </} On SO NO oo 00 q SO rsi 00 OS CO

rsi ci no oo i> ON ts in SO SO rn oo 00 fsj

i ro

i

<x|

i r~i1 1

i i

1

1

i

ox s=©s N=

os ox

S=> 0s s=ox N=

ox

s?cS s? s?OS s?OS s?oS s?OS s=OS s?

os s?OS s?©s so

OS

sO

ox s5OS 6s- "* oo oo oo 00 oo 00 OO r- >0 <s) 00 00 O CO

\6 d co co CO tt rt d rn <sj csj oo" st

CN

1 i jo1 r>]1 i 1 l 1 l i 1 fsii fsj1

(*s)

1 1 CslI fsl1 fsl1 Csli 1

N©

ox X=ox

no

os

SOox

s°oN

s§ S=©S s? s? s?OS s?OS

vO0s" s?

os

s^©s s?OS s?OS s?OS s?

os

soos

s°OS N° 00 Csl OO r—

.

[3 Q

C\ oo i

1 00 ro CO rn r~ d <sj 00 l-H

. o d CO ci Os d fN

i i roi <Nl coi COl tsi fSi ts|1

<s|

1 i m cs CO1 mi mi mi i rsii CO1

-* r~; •st r- r- r- m oo SO p SO OS m CSl OS

t~ o IN os r> d rn <sj d fsi oo Os SO iS 00

OS in so o CN cs r- (si ON lO sri o oo 00 oo

<sl OO sr> ON m OS OS o sn Os m o t p ro m

©S s?©s N°OS s?OS

\°©s s?©s s?OS s?oS ©ssf s?OS 00 p CSj o P co p csi P oi iri Csj 00 OS oo m m K-S VI CO rs rs CS

csloOsOOsOsoooo--Tf

Tfrorococorococococo

os o — oo d d oo ro

rsi ^oors)^«osco«vs^-r~;iov-ioosq dK — (sid^CsjrtcSOsrO—^r^dos csl —< — ro ~

s?

OS

vs

SO

CO

oS SOOS s?oS s=OS s?OS S=

ox s=

os

S<= OS s?OS s?OS sPOS s?ox

SO©s SOoS s?OS s=OS s=OS s?

os

soOS s=©S s=©S

OS rsi t-~- ri CO co v> co CO ro CO Os 00 P SO OS p p d d rsi P >* oo oo OO OO CO OS rsi csi d CO co ro CO CO co (S| rs rs) «s) rsi CO CO rsi CO ro CO co co CO CO

csj oo

r--

— (NroTtsnsor-ooos oH

104
SESSION 3

WIND ENGINEERING

Aerodynamic Stability of a Super Long-Span Bridge with Slotted Box Girder

by

Hiroshi Sato

1

, Nobuyuki Hirahara2

, Koichiro Fumoto3

, Shigeru Hirano4 and Shigeki Kusuhara5

ABSTRACT

Aerodynamic stability is one of the most

important themes in the design of super

long-span bridges. In order to improve the

aerodynamic stability, various researches were

conducted and it was confirmed that slotted box

girder with some devices shows the good

aerodynamic stability. In this paper, the results

of full aeroelastic model test and flutter analysis

for a generic super long-span bridge, whose

main span is 2.8km, are described. And it was

confirmed that slotted box girder was applicable

for the super long-span bridge from the reason

that it is excellent in economical efficiency and

aerodynamic stability.

KEYWORDS: aerodynamic stability; super

long-span bridge; flutter analysis; full

aeroelastic model; slotted box girder

1. INTRODUCTION

There are several plans or ideas of strait crossing

road projects in Japan. In these projects, super

long-span bridges longer than the Akashi

Kaikyo Bridge may be included. In the design of

such super long-span bridges, aerodynamic

stability is one of the most important themes.

Furthermore, reduction of construction cost is also required. Therefore, Independent

Administrative Institution Public Works

Research Institute (IAIPWRI) and

Honshu-Shikoku Bridge Authority (HSBA)

have been conducting cooperative study on

super structure of super long-span bridges that

have good aerodynamic stability and economical

efficiency.

In the previous studies [1][2] by the authors, it was found that a slotted box girder with some

devices has good aerodynamic characteristics.

Therefore, the slotted box girder was applied to a generic super long-span bridge, and full

aeroelastic model test was conducted. And the

result was compared with 3-dimensional flutter

analysis.

In this paper, main results of the previous study

on the slotted box girders are outlined first. Then

the results of full aeroelastic model test and

flutter analysis for a generic super long-span

bridge, whose main span is 2.8km, are described.

2. FLUTTER CHARACTERISTICS OF

SLOTTED BOX GIRDERS [1]

The effect of location and size of slot on

aerodynamic characteristics was examined

through section model wind tunnel tests. Reduced mass fl (=m/( p B

2

), m: mass per unit

1 Director, Structures Research Group, Public Works Research Institute, Independent Administrative

Institution

2Manager, Maintenance Planning Division, Maintenance Department, Honshu-Shikoku Bridge Authority

(Formerly, Leader of Bridge Structure Research Team, Structures Research Group, ditto)

3Senior Research Engineer, Bridge Structure Research Team, ditto

4Director, Tarumi Maintenance Work Office, The 1st Maintenance Bureau, Honshu-Shikoku Bridge

Authority (Formerly, Manager, Engineering Management Division, Long-span Bridge Engineering

Center, HSBA)

5 Deputy Manager, Engineering Management Division, ditto

107
length, p : air density, B: girder width), reduced

polar moment of inertia V (=I/( p B

4

), I: polar

moment of inertia per unit length), and natural

frequency ratio £ (=fe/fz, f e: torsional natural

frequency, fz: vertical bending natural

frequency) were 16, 2.1, and 2.1, respectively.

The cross section of the model is shown in Fig. 1

.

From the test results, it was found that the slot at the center increased the flutter onset wind speed.

It was also found that the flutter onset wind

speed was increased with the width of slot at the

center of the girder (Fig.2).

In order to understand the effect of slot at the

center of girder, preliminary analysis was

conducted for slotted plate. For the analysis,

aerodynamic forces acting on each plate was

calculated using the Theodorsen's function. The

aerodynamic interference between the 2 boxes

was neglected. Using these aerodynamic forces,

two degree-of-freedom flutter analysis was

conducted by U-g method. The result of the

flutter analysis (Fig.2) indicated that the flutter

onset wind speed increased with size of slot. The

differences between the analysis and the

experiment seemed to be caused by

aerodynamic interference between the 2 boxes.

Although wide slot at the center of the girder

improves flutter characteristics, narrower slot

would be preferable from the viewpoint of

construction cost of towers and foundations. To

improve aerodynamic characteristics, the effect

of some devices was studied by section model

tests. The tested devices are illustrated in Fig.3.

The results showed that the center barrier and

guide vanes improved flutter characteristics very

well (Fig.4). However, the flutter speed was not

so high when angle of attack was -3 deg. It was

found that the guard rails at the bottom deck

increased the flutter speed considerably at this

angle of attack (Fig.5).

3. UNSTEADY AERODYNAMIC FORCES

OF SLOTTED BOX GIRDERS [2]

From the above studies, it was found that slotted

plates and slotted box girders have better flutter

characteristics than single plates and single box

girders. It was also found that the devices such

as center barrier and guide vanes are effective to

improve flutter characteristics of slotted box

girders.

In order to understand causes of flutter

characteristics of slotted plates and box girders

more precisely, unsteady aerodynamic forces

were measured for three models: model A

(single box girder, b=0 in Fig.l), model B

(slotted box girder, b=0.22B in Fig.l) and model

C (slotted box girder with devices, Fig.3). The

measurement was made by forced oscillation

method with angle of attack 0 degree.

Coefficients of the unsteady aerodynamic forces

were defined as follows:

L=7T p {B2 [LZR a)

2z+Lzia)z']

+B3 [L eR aj

2 0+L ei O)0']} (l.l)

M=7T p {B3 [MZR OJ

2z+MZI aJz']

+B4[M eR OJ

2 0+Mei w 0']} (1.2)

where, L: lift (upward positive), M:

aerodynamic moment (head up positive), z:

vertical displacement (upward positive), 0 :

torsional displacement (head up positive), (x): circular frequency, (

)': d( )/dt, Lxx or Mxx : coefficients of unsteady aerodynamic forces (z : caused by vertical vibration, e

: caused by

torsional vibration, R : in phase with

displacement, t

: in phase with velocity).

In general, it is difficult to predict coupled

flutter characteristics directly from these

coefficients. For 2-degrees of freedom system,

Nakamura [3] showed approximate relationship

between unsteady aerodynamic coefficients MZI , M0 i, L eR and M 0R and some flutter properties

as follows:

8 a=- 71

2MZIX/ V - 7T

2Meil V

(2.1)

X =Zo/0o/B = 7rL eR/(-l+(fz/f e )

2 C72)/// (2.2)

a

2 =(fe /f)2 = i+7rM eR/y (2.3)

where, 8 a: aerodynamic damping in

logarithmic decrement.

108
If onset of flutter is defined as 8 a^O, simpler

condition for onset of flutter can be derived

from (2.1)-(2.3) as follows:

OfMz, L 9R/M ei+jSM flR ^l (3.1)

a=(e 2/(e

2-l))(n/u) (3.2)

£=(l/(£ 2-l))(7T/i/) (3.3)

The left hand side of inequality (3.1) was

calculated for the Models A, B and C using

measured unsteady aerodynamic forces, as well

as for single plate and slotted plate using the

Theodorsen's function. £ , U and V were

assumed as 2.0, 15 and 2.0, respectively. The

results are shown in Fig.6. The slotted box

girders and slotted plate show higher flutter

speed than the single box girder or single plate.

Since the first term of the left hand side of

inequality (3.1) is much larger than the second

term, it can be said that this higher flutter speed

was caused mainly by the property of MZi L e

In Fig.6, reduced flutter speed U/(fB) of slotted

box girder with devices is almost identical with

that of slotted box girder. In Fig.7, the results

are plotted with f e B/U. Flutter speed of slotted

box girder with devices is higher than that

without devices. It means that the effect of

devices came from small value of MeR, which

affected apparent frequency in wind as was

shown in Equation (2.3).

4. WIND TUNNEL STUDY WITH A

FULL AEROELASTIC MODEL

4.1 Full aeroelastic model

A suspension bridge, whose main span length is 2,800m and total length is 5,000m (Fig.8), was

assumed as a prototype bridge of this study. The

cross section of stiffening girder is shown in

Fig.9. This box girder has 4-lanes, whose

2-lanes are on the grating installed on the slot in

consideration of economical efficiency.

Wind tunnel test was conducted at the large

boundary wind tunnel facility in Tsukuba. It was

built in 1991 as one of the cooperative efforts

between PWRI and HSBA in order to verify the

aerodynamic stability of super long-span bridge

such as the Akashi Kaikyo Bridge, and to establish the wind resistant design method for

super long-span bridges considering

3-dimentional effects of structure and flow. The

test section of this wind tunnel is 41m wide, 4m

high and 30m long, and maximum wind speed is 12m/s. It is one of the largest boundary layer

wind tunnels in the world.

The scale ratio of a full aeroelastic model of the

assumed super long-span bridge was 1/125.

Structural dimensions of the bridge and the

model are shown in Table 1

.

4.2 Wind tunnel test results for the full

aeroelastic model

Static deformation by wind load is shown in

Fig. 10 and Photo 1. Large horizontal

deformation (leeward side) and large torsional

deformation (head down) were measured at the

center of main span.

Logarithmic decrement at each wind speed is shown in Fig.ll. It increased gradually with

wind speed, and began decreasing at wind speed

of about 6m/s. After that, logarithmic decrement

decreased, and changed to negative value at 8.8m/s, and flutter started. According to the

similarity law of Froude's number, flutter speed

for the assumed bridge would be about lOOm/s.

5. FLUTTER ANALYSIS FOR THE

FULL AEROELASTIC MODEL

3 -dimensional flutter analysis for the full

aeroelastic model was conducted by using the

measured unsteady aerodynamic forces. The

analytical method was the same as the one that

was used for the Akashi Kaikyo Bridge [4]. The

conditions of 3 -dimensional flutter analysis are shown in Table 2. In the analysis, static

deformation was calculated first, then unsteady

aerodynamic forces corresponding to attack

angle of the girder were introduced. After that,

eigenvalues were calculated using 50 vibrational

modes in still air. Modified torsional stiffness

was used for the calculation so that torsional

natural frequency might become equal to the

measured value.

The analytical results of static deformation were

109
shown in Fig. 12. As for horizontal displacement

and vertical displacement, the analytical values

agreed well with experimental values. However,

the analytical value (absolute value) for the

torsional deformation was a little smaller than

experimental value. Since aerodynamic

characteristics are sensitive to attack angles, the

flutter analysis was carried out using the

measured torsional deformation.

The change of apparent damping of the 1st

torsional symmetrical mode (mode 19) is shown

in Fig. 13 by triangle mark. The flutter speed

from the 3 -dimensional flutter analysis was

7.9m/s. It agreed fairly well with the

experiment.

Since the mass and polar moment of inertia of

the model were larger than the values required

from the similarity laws, flutter analysis was

conducted for the required values of the model.

The flutter speed was 7.6m/s. Therefore, flutter

speed for the assumed bridge would be

estimated about 85m/s, which confirmed its aeroelastic stability

REFERENCES:

[1] Sato H. and Ogihara K., Aerodynamic

characteristics of slotted box girders,

Proceedings of the 28th Joint Meeting of the

Panel on Wind and Seismic Effects, UJNR,

1996

[2] Sato H., Ogihara K. and Ogi K.,

Consideration on flutter characteristics of

super long-span bridges, Proceedings of the

30th Joint Meeting of the Panel on Wind

and Seismic Effects, UJNR, 1998

[3] Nakamura, Y., An analysis of binary flutter

of bridge deck sections, J. of Sound and

Vibration, 57(4), 1978

[4] Miyata T., Sato H. et al., Full model wind

tunnel study on the Akashi Kaikyo Bridge,

Proceedings of the 9th International

Conference on Wind Engineering, 1995,

pp.793-798

6. CONCLUSIONS

Full aeroelastic model test and 3-dimentional

flutter analysis was conducted for a generic

super long-span bridge. Results and conclusions

obtained are summarized as follows:

1) The slotted box girder was applied to a generic super long-span bridge, whose main span

length was assumed to be 2.8km. Wind tunnel

study was conducted with a 1/125 full

aeroelastic model in smooth flow. It was con- firmed that its flutter speed was high enough.

2) Three-dimensional flutter analysis was

conducted for the full aeroelastic model. In the

analysis, unsteady aerodynamic forces corre- sponding to attack angles of the girder deformed

by steady wind forces were introduced. The

calculated flutter speed agreed fairly well with

the experiment.

In order to predict the torsional deformation and

the flutter speed more accurately, improvement

of the analytical model for the bridge is in

progress.

110
B

0.051B

< JO-= B/17

Fig.l Cross Section of Slotted Girder

25

0

1

1

1

1

1

A

1

1

1

1

[

Analysis

A Experiment

0.0 0.1 0.2 0.3

Slot Ratio b/B

0.4 0.5

Fig. 2 Flutter Onset Speed and Slot Ratio

Guard Rails

Guard Rails

Center Barrier

1

Guide Vane

b=0.22B k — H Guide Vane

B

d=B/17

0.051B

Fig.3 Slotted Box Girder with Devices

111
2.0

1.5

o

5

1.0 -

0.5 -

0.0

> < >

Angle of Attack

Basic

Fig.4 Effect of Tested Devices

Center Barrier Center Barrier

+Guide Vane

2.0

1.5 -

1.0 -

0.5 -

0.0

t

• Without Guard Rails

With Guard Rails

No flutter was observed

up to the wind speed

-3 0 3

Angle of Attack

Fig.5 Effect of Guard Rails at Bottom Deck

112
3 2

U 1

a 0

0.01 _ 0.1 Reduced Frequency fBAJ

(a) Single Plate

OS

S3I—

I

Is

0

X

%

1\

- >

-

1

0.01 0.1 Reduced Frequency fBAJ

(b) Slotted plate

0.01 0.1 1 Reduced Frequency fBAJ

(c) Single Box Girder

0.01 0.1 1 Reduced Frequency fBAJ

(d) Slotted Box Girder

*3

S3I—

i

a

0

0.01 0.1 1 Reduced Frequency fBAJ

:ctMziLeR/M9i + pMeR

A :aMziLeB/Mei

(d) Slotted Box Girder with Devices

Fig.6 Prediction of Flutter Onset (plotted with fBAJ)

[ Flutter will take place when :ctMziLeR/M9i + (3MeR ^ 1 ]

113
— single plate slotted plate

X

single box girder A slotted box girder

|-| slotted box girder with devices

Fig.7 Prediction of Flutter Onset (plotted with feB/U)

[ Flutter will take place when :aMziLeR/Mei + PM6r ^ 1 ]

114
Fig. 8 Assumed super long-span bridge

unit:*

Cable span 28.0m

grating

Openning width 10.3m

Total width 34.5m

. guide vane

Fig. 9 Cross section of slotted box girder

115
V:\viiid velocity

01 : inclination of girder

10 20 30 40 50

Horizontal displacernent(cm)

Fig. 1 0 Deformation of girder

0.3

0 2 4 6 8 10

Wind velocity in wind tunnel (nVs)

Fig. 1 1 Apparent damping - wind velocity

116
Experiment • Analysis

10 Wind velocity in wind tunnel (m/s)

Experiment

# Analysis

10 Wind velocity in wind tunnel (m/s)

Experiment • Analysis

Wind velocity in wind tunnel (m/s)

10

Fig. 12 Results of static deformation

117
0.30

0.20

I

•8 0.10

u§

3

$ 0.00

0.

-0.10

A Analysis

• Experiment

A A \

8.8m/s

0 1.0 2.0 3.0 4.0 5.0 JUJL 7. 7.9m/s

•

0

Wind velocity in wind tunnel (m/s)

Fig. 13 Apparent damping - wind velocity (1st torsion symmetrical mode)

Photo 1 Static deformation of super long-span bridge (wind speed=8m/s)

118
Table 1 Structure dimensions of super long-span bridge

Assumed bridge

Model value

(b)/(a)

Required (a) Measured (b)

Scale - 1/125 1/1 1/175 - Mass M 28.41t/m 1 8181ce/m 1 100kp/m 1.156

Polar moment of inertia Id 388t-m2/m H HOI ^QVn • m2 /m u.uuz'toKg m /m 1.547

Girder size Width B 34.5m U.Z- / QUI ft 97fim 1.000

Depth D 4.0m 0 032m 1.000

Stiffness

Vertical Ely 1.707kNm2/box 9 384Nm2/box 1 2 277Nm2/box 1.308

Horizontal EIH 3.107kNm2/box 48 828Nm2 Tt) .O^Ol ^ 111 /hnx /UUA 1 2 277Nm2

1 f / 1^1 111 /UUA /hmc 0.251

Torsional GJ 1.090kNm2/box Q06Nm2/hf»Y fi 853Nm2/hr>Y 0.992

Natural

frequency

Vertical

1

st 0.0621Hz U.U7J1 VL, 0 674H7 0.973

2

nd 0.0990Hz 1.111Hz 1.106Hz 0.995

[1st] 0.0610Hz 0.683Hz 0.703Hz 1.029

[2nd] 0.0831Hz 1.499Hz 1.494Hz 0.997

Torsional

1

st 0.1256Hz 1.369Hz 1.321Hz 0.965

[1st] 0.1466Hz 1.600Hz 1.597Hz 0.998

Horizontal

1

st 0.0316Hz 0.352Hz 0.352Hz 1.000

[1st] 0.0550Hz 0.502Hz 0.491Hz 0.978

[ ] means asymmetrical mode

Table 2 The conditions of flutter analysis

Item Analysis condition

Analytical

method

Mode combination method.

Using lower 50 modes.

Air density 0.12 kg/m3

Structural

damping All modes 6=0.02

Main girder

Coefficients

of

aerodynami

c forces

Foree"""""""""" ^ Vertical Torsional Horizontal

Lift O O A

Moment • O O A

Drag O O A

OjUnsteady aerodynamic forces

A;Quasi-steady aerodynamic forces

Cable

Quasi-steady drag and lift force (Cd=0.7)

Tower

Not considered

119

Wind Effects on Long Span Cable Stayed Bridges: Assessment and Validation

by

Nicholas P. Jones

1 and Ender Ozkan2

ABSTRACT

The well-known collapse of Tacoma Narrows

Bridge in 1 940 clearly identified the importance of

aeroelastic effects on long-span bridge

performance. Extensive research has been carried

out since then to better understand the effects of

wind on long-span bridges, producing various

analytical response prediction techniques. An

example ofthe application of such techniques will

be presented. However, due to challenges related

with full-scale measurements, these prediction

techniques have commonly been validated using

only wind-tunnel experiments. Recent research has

revolved around the conduct of long-term fullscale measurements on a cable-stayed bridge to

compare actual bridge performance with those of

analytical predictions. In order to ensure the

reliability of predicted response, the input

parameters, such as wind conditions at the site and

modal properties of the bridge are also calibrated

using corresponding measured quantities. This

paper will also summarize some ofthe preliminary

results and outline their implications.

KEYWORDS: bridge engineering, wind

engineering, flutter, buffeting, long-span bridges

1.0 INTRODUCTION

The maximum span of long-span bridges has been

extended in recent decades to where today, the

Akashi-Kaikyo Bridge (central span 1991 m) has

been completed, and longer bridges are planned

(e.g., Messina Straits (3300 m), Gibraltar Straits

(5000m)). These successes are due in particular to

progress in wind-resistant design; a primary

component in the design of long-span bridges.

Recently, multi-mode flutter and buffeting analysis

procedures have been developed. These

procedures, which were based on frequency-domain

methods, take into account the fully coupled

aeroelastic and aerodynamic response of long-span

bridges to wind excitation.

The current methodology for the estimation of the

response of long-span bridges to wind loads

incorporates the following components:

• Measurement of a comprehensive set of

aerodynamic and aeroelastic parameters for a given cross section using a suitably (i.e.,

aerodynamically) scaled section model.

These parameters include: the static

coefficients (lift, moment, and drag at a number of different angles of attack) and the

flutter derivatives, generally also at several

positive and negative angles of attack. It is emphasized that these quantities are intended

to be sectional quantities that will be used in

the analytical model. Examples of procedures can be found in Sarkar et al. (1994) and Singh

et al (1996).

• A detailed numerical (generally finite

element) dynamic model for the bridge under

consideration. This model will be expected to

provide a set of eigenvalues and eigenvectors

for the structure and a corresponding set of

generalized inertias. Generally, this will

include at least 20 modes, but in some cases more may be required (e.g., up to 50 for very

long structures).

• An analytical framework and computational

aids for synthesizing the above data. Scanlan

and Jones (1990) provide a comprehensive

overview of the single-degree of freedom

theory, while Jain et al. (1996a, 1996b) and

Katsuchi et al. (1998a, 1998b, 1999) outline

the extension of this theory to consider the

interaction of multiple modes.

In addition to the above components, knowledge

Professor and Chair, Dept. of Civil Engineering, Johns Hopkins University, Baltimore, MD 21218, USA

Graduate student, Dept. of Civil Engineering, Johns Hopkins University, Baltimore, MD 21218, USA

121
or assumptions about modal damping values and

the wind environment are also required.

One of the most commonly used prediction

techniques is the frequency-domain analysis

described by Scanlan and Jones (1990), Jain et al. (1996a) and Katsuchi et al. (1999), and is based on

modal analysis in the frequency domain. A suite of

computer programs has been created to enable fast

and efficient implementation of multi-mode

frequency-domain prediction methods. For brevity,

details are not included here; interested readers are referred to the above-referenced publications for

this material. —

2.0 EXAMPLE APPLICATION

This section summarizes the results of the flutter

and buffeting analysis carried out for a cablestayed bridge: a 1 169 ft. main span structure and

shows the capabilities of frequency-domain-based

analysis tools. The input parameters in Table 1 were used in the analysis to model the structural

and climatological properties of the bridge. Table

2 lists this first ten modal frequencies ofthe bridge,

and their associated descriptions. The A*2 flutter

derivative for different wind directions and angles

of attack are shown in Figure 1. The potential

flutter susceptibility is indicated by the crossing of

the A*2 from negative to positive at a reduced

velocity (U/nB) of about 4.

2.1 Flutter analysis

For some long-span bridges, the lowest torsional

mode may couple with the first vertical mode of

the bridge to cause flutter at a lower wind velocity

than predicted from the single-mode analysis

alone. To evaluate the possible coupling ofmodes,

a two-mode flutter analysis was carried out for the

bridge using Modes 2 and 7 (first vertical and first torsional). The two-mode flutter analysis was

repeated using all of the flutter derivative sets; results are summarized in Table 3.

According to these results, the most critical

condition corresponds to a south with an angle of

attack of -3 degrees. Coupling was observed to be

minimal in this case: this is primarily a singlemode flutter, as would generally be expected for a

structure of this size.

2.2 Buffeting Analysis

Buffeting analysis of the structure was carried out

using two sets of flutter derivative data, one with

force coefficients corresponding to a north wind at zero degrees) and one with force coefficients

corresponding to a south wind at zero degrees. In

the results reported herein, the mean wind speed U

is taken as 33 m/s (109 ft/s).

Table 3 summarizes the results ofthis investigation,

and Figures 2 and 3 show the RMS vertical and

torsional displacement estimated as a function of

span location for varying number ofmodes included

in the analysis. Note that the responses converge

rapidly after the significant modes that contribute

are included in the analysis. Figures 4 and 5 show

spectra of response at midspan and quarter-span

points, and demonstrate the different modal

contributions expected at these locations. Again,

analysis confirmed that the modal coupling in this

structure is insignificant, and analysis using a modeby-mode approach is generally adequate.

3.0 FULL-SCALE MEASUREMENTS

As noted above, few opportunities have existed to quantitatively verify the above results at full scale.

A long-term measurement program is currently

underway at the Fred Hartman Bridge in Texas to continuously monitor its response to ambient

loading conditions. The Fred Hartman Bridge (Fig.

6) is a twin-deck cable-stayed bridge with a main

span of 380 m. and two side spans of 147 m. The

decks are carried by a total of 192 cables, arranged

in four inclined planes and connected to the deck at 15 m intervals.

Measurements are carried out using a PC-based self- triggering system that collects data on the basis of

exceedance of threshold motion and wind levels.

This system has been in use for over 4 years,

collecting in excess of20000 trigger files. Each file contains 5-minute trigger runs collected at 40 Hz,

which are stored on high-capacity disks. The

recorded files are later processed to extract deck and

122
cable accelerations, cable displacements and loads,

wind speeds and other relevant meteorological

factors, all of which are summarized in a comprehensive database for subsequent analysis.

Further details of data collection and processing can be found in Ozkan et al. (2001a).

As noted earlier, there are three main components

used in a bridge wind analysis: knowledge of the

meteorological conditions at the bridge site,

information on the modal properties of the

structure and the experimentally determined flutter

derivatives and associated aerodynamic parameters

associated with the cross section. The first component, through the wind spectra, can be

evaluated directly using measured wind speeds and

compared to theoretical meteorological models.

Preliminary comparison of wind spectra show

reasonable agreement (see Figure 7); detailed

results will not be reported herein.

The modal properties predicted for the bridge can

also be evaluated directly from the measured deck

frequencies through comparison to the results

obtained from a 3-D finite element model of the

bridge; good agreement has generally been

obtained. It should be noted here that due to the

inherent flexibility of long-span bridges, selfexcited forces play a role in the overall stiffness

and damping of the structure, making them windspeed dependent. This characteristic is modeled

through the flutter derivatives, the effects ofwhich

can be seen in histograms of modal frequencies,

which show a range of values for each mode.

Studies are also being performed to better evaluate

the wind-speed dependence of frequency and

damping of the structure and relate them to the

flutter derivatives.

Comparison of overall bridge response is made

using root-mean-square (RMS) acceleration and

displacement response of the bridge. Since

multiple modes contribute to the response, RMS

displacement is estimated for each mode

individually. This is done for each record, and the

results consequently compared to analytical

predictions obtained from the programs mentioned

above. However, the nature ofthe structure and the

measurement program gives rise to observations

that should be considered carefully to ensure the

reliability of results. An example of these is the

localized effect of cable vibrations. These are instances of large-amplitude cable vibrations that are recorded by deck transducers (Ozkan et al. 200 1 a). Measurements taken after the installation of

dampers on the cables show a much-reduced level

of such vibrations and more evident global

vibrations of the deck. Another interesting

observation is the instances of cable-deck

interaction, initiated by moderate vibrations of the

deck that induces oscillation of cables (Ozkan et al. 2001b).

3.1 Deck Vibration Measurements and Data

Analysis

A significant challenge related with long-term

monitoring projects is the process ofanalyzing large

quantities of data without loss of interesting

information that might not have been expected.

More than 20000 trigger files have been recorded

during the three years of the test program and more

are continuously being collected. Analyzing such a

large number of data files demands extensive use of

automated procedures. However, the use of such

procedures must be carefully controlled to ensure

that flawed or questionable data are not included in

the study. The data analysis techniques used in this

project have been automated to the greatest extent

considered prudent, with careful consideration to maintaining the integrity and accuracy of the data

and not missing important features of individual

records.

The recorded files are initially processed to determine the general features of the raw data.

These features include the mean, standard deviation

and other higher moments of the data as well as one-minute average wind speeds and directions,

accelerations and displacements, all of which are automatically added to a database. The database

provides for easy analysis and correlation of these

statistical quantities. It is also possible to readily

interrogate the data using queries created and stored

in the database.

Modal frequencies and mode shapes have been

found and these values have been compared with

values obtained from finite element (FE) analysis.

Preliminary results of this comparison have been

123
presented previously (Ozkan et al. 2001a). Table 5

shows the comparison of measured modal

frequencies with those obtained from a FE analysis

for the first 20 modes. In general, good agreement

between the two data sets is observed. Similarly,

the mode shapes have been found for the first 20

modes, showing reasonable agreement with those

calculated from the FE analysis (Ozkan et al. 2001a).

To investigate the wind-vibration characteristics of

the bridge deck, plots of root mean square (RMS)

displacement of the deck versus wind speed have

been made (Figure 8). A general trend of

increasing RMS acceleration with wind speed can be observed from this figure. Comparison with

predicted responses using the procedure outlined

above and parameters suitable for the recording

environment show reasonable agreement in large

measure; "outliers" are currently being studied in

more detail.

3.2 Deck-Stay Interaction

An example of an interesting record is given in

Fig. 9. This record forms the initial five-minutes of

a series of triggers during the passage of a

meteorological event. For this specific record, Fig.

9(a) and (b) show the time-histories and power

spectral densities (PSD's) of vertical deck

acceleration at midspan and of the adjacent stay

cable AS24 (length 198 m; natural frequency

approximately 0.59Hz), respectively. Fig. 9(c)

shows the wind speed at deck level. These figures,

and others similar that have been made for

different deck instruments, show a dominant

frequency of vibration at approximately 0.58 Hz.

It is important to note that this frequency

corresponds to the third symmetric vertical mode

of the deck (given as approximately 0.56 Hz in

Table- 1), and is also close to the first mode of the

stay cable AS24. This is an interesting and

important observation since the first-mode

vibrations of a cable at this level ofacceleration are generally associated with large displacements. In

fact, by integrating the acceleration time-history, a displacement amplitude of approximately one

meter (peak to peak) was estimated.

Furthermore, by observing the time-histories it can

be seen that the significant vibrations are initially

observed at the deck instead of the cable. This

observation, as well as the similarity of modal

frequencies suggests that the deck is driving the

cable to vibrate with large amplitude in its fundamental mode. Vortex-induced vibration ofthe

deck is thought to be the driving mechanism for this

motion. Further studies are continuing to better

understand the underlying mechanisms involved in

this behavior and its consequences, and will be

reported in future publications.

4.0 CONCLUDING REMARKS

The preceding paper presents an overview of longterm efforts to monitor a cable-stayed bridge for a

variety of purposes, including understanding the

modal characteristics and wind-induced responses

under ambient wind conditions. Using data files

collected during various meteorological conditions,

natural frequencies and mode shapes of the deck were found using an automated data analysis

procedure. The measured modal frequencies, mode

shapes, and RMS responses are observed to agree

with the predicted values found from a finite

element and aerodynamic analysis.

The analysis and comparison described above

demonstrates acceptable performance of the

analytical model in its ability to predict the response

of long-span bridges to wind loading. The

analytical model (with suitable choice of

parameters) may therefore be considered suitable for

the prediction of the prototype response. Overall,

the methodology employed in the present paper

represents the general versatility of the analytical

and experimental techniques for the aeroelastic

design of long-span bridges.

It is noted that while automated data analysis

procedures are clearly necessary for such large

volumes of data, care must be taken not to miss or obscure important phenomena or characteristics

through this approach. The interaction observed

between the deck and a stay is a good example of a

situation where careful interpretation of the data is required to fully understand the relevant underlying

mechanics.

5.0 ACKNOWLEDGEMENTS

124
The authors would like to thank Dr. Jon Raggett of

West Wind Lab, Carmel, CA for performing the

aerodynamic and aeroelastic section model tests reported in this paper. URS Corporation was

responsible for the finite element analysis of the

first structure analyzed herein that provided the

structural dynamic data.

The support of both authors by the National

Science Foundation is also gratefully

acknowledged. The authors would also like to acknowledge the Texas Department of

Transportation, the US Federal Highway

Administration, the US Department of Defense

(NDSEG Fellowship) and the Department of Civil

Engineering at Johns Hopkins University for the

support of various components of this work.

6.0 REFERENCES

Jain, A., Jones, N.P. and Scanlan, R.H. (1996a).

"Coupled Flutter and Buffeting Analysis of LongSpan Bridges." J. Str. Engrg., ASCE. 122(7), 716-

725.

Jain, A., Jones, N.P., and Scanlan, R.H. "Coupled

Aeroelastic and Aerodynamic Analysis of

Long-Span Bridges." J. Wind Eng. and Indust.

Aero, 60 (1996b), 69-80.

Jain, A., Jones, N.P., and Scanlan, R.H. "Effect of

Modal Damping on Bridge Aeroelasticity." J. Wind Eng. and Indust. Aero, 77&78, 421-430,

(1998).

Katsuchi, H., Jones, N. P., Scanlan, R. H. and

Akiyama, H. (1998a). "A Study ofMode Coupling

in Flutter and Buffeting of the Akashi-Kaikyo

Bridge." J. Struct. Mech. and Earthquake

Engrg. , JSCE, No 598 / 1-44.

Katsuchi, H., Jones, N. P., Scanlan, R. H. and

Akiyama, H. "Multi-Mode Flutter and Buffeting

Analysis of the Akashi-Kaikyo Bridge." J. Wind

Eng. and Indust. Aero, 77&78, 431-441, (1998b).

Katsuchi, H., Jones, N.P., Scanlan, R.H., and

Akiyama, H. "Multimode Coupled Buffeting and

Flutter Analysis of the Akashi-Kaikyo Bridge". J.

Struct. Eng. ASCE, 125(1), 60-70, January, 1999.

Ozkan, E., Main, J. and Jones, N. P. (2001a).

"Long-Term Measurements on a Cable Stayed

Bridge", Proc. 19th Inter. Modal Analysis Conf.,

Kissimmee, Florida, p. 48-50.

Ozkan, E., Main, J. and Jones, N. P. (2001b). "FullScale Measurements on the Fred Hartman Bridge",

Proc. 5

th Asia-Pacific Conf. on WindEngrg., Kyoto,

Japan.

Sarkar, P.P., Jones, N.P. and Scanlan RH. "A

Recursive Time-Domain System-Identification

Procedure for Extraction of Aeroelastic

Parameters." Journal of Engineering Mechanics,

120(8), ASCE, August, 1994, 1718-1742.

Scanlan R.H. and Jones N.P. "On Aerodynamic

Admittance in Bridge Aeroelastic Analysis." J. Fluids and Structures, 13, 1017-1027 (1999)

Scanlan, R.H. and Jones, N.P. (1990). "Aeroelastic

Analysis of Cable-Stayed Bridges." J. Struct. Engrg., ASCE, 116(2), 279-297.

Scanlan, R.H.. "State-of-the-Art Methods for

Calculating Flutter, Vortex-Induced, and Buffeting

Response of Bridge Structures." Federal Highway

Administration Report No. FHWA/RD-80/50

(1981).

Simiu, E. and Scanlan, R.H., Wind Effects on

Structures, Third Edition, John Wiley and Sons,

New York, N.Y., 1996

Singh, L., Jones, N.P., Scanlan, R.H. and

Lorendeaux, O. "Identification of Lateral Flutter

Derivatives of Long-Span Bridges." J. Wind Eng.

and Indust. Aero, 60 (1996), 81-89.

125
Table 1 : Parameters for Examp e Structure

C (structural damping) 0.003

B (width) - ft 76.75

L (total length) - ft 1169.0

Air Density, o. - lbs

2/ft

4 0.002378

Lift coeffcient Cl -U.Uj44

Drag coeffcient Co 0.134

Moment coeffcient Cm -0.0293

Denvative 01 lift coetiicient CL

J. 1U/

Derivative of moment coefficient CM

0.332

Z (deck height) - ft 78

Z0 (roughness length) - ft 0.25

Correlation Constant 5

Table 2: Modal frequencies

Mode Frequency Period

No. (Hz.) (Second) Description

1 0.172 5.825 1st Longitudinal Deck Mode

2 0.294 3.398 1st Vertical Deck Mode

3 0.446 2.241 1 st Lateral Deck Mode

4 0.473 2.113

5 0.718 1.392

6 0.723 1 .383 1 st Torsional Deck Mode

7 0.784 1.276

8 0.867 1.153

9 0.919 1.088

10 0.933 1 .072

Table 3: Critical flutter velocities

Angle of Critical Flutter

Attack Velocity

North 0 deg 163 mph

North +3 deg >191 mph

North -3 deg 158 mph

South -3 deg 145 mph

South 0 deg 161 mph

South +3 deg > 178 mph

126
Table 4: Buffeting analysis results (0.5% damping)

Configuration Modes Analyzed

Vertical RMS Disp. at

IVlUJLbJJall ^111^11 ^vlli^ j

North Wind

1 Strl A AA (\ 1 X\

1 9 & % t.*tJ ^ 1 L .J J

1 \J lvlUUCS A 48 n 1 41

18 Modes 4.49(11.4)

South Wind

1 &2 4.52(11.5)

1, 2 & 3 4.53(11.5)

Table 5: Comparison of measured deck modes with FE analysis

T .one-Term ii Villi Phasins

Measured FEM Percentage ( I :in-phase Description

Mode Frequency (Hz) T \ L Frequency (Hz) Difference O: Out-of-phase) of the Mode (FE)

1 0.290 0.286 1.4 I Vertical

2 0.299 0.291 2.8 O Vertical

3 0.375 0.366 2.5 I Vertical

4 0.377 O Vertical

5 0.432 0.410 5.4 0 Lateral

6 0.426 I Lateral

7 0.564 0.556 1.4 I Vertical

8 0.562 0 Vertical

9 0.612 0 Torsional

10 0.586 0.625 6.2 I Vertical

11 0.634 0 Vertical

12 0.665 0.658 1.1 I Vertical

13 0.659 Torsional-Lateral

14 0.683 0.662 3.2 Torsional-Bending

15 0.714 0.735 2.9 I Vertical

16 0.736 o Vertical

17 0.756 I Torsional

18 0.784 0.757 3.6 0 Vertical

19 0.817 I Torsional

20 0.924 0.856 7.9 I Vertical

127
0.25

-0.25

-0.5

U/nB

Figure 1: Flutter derivative A*2 for various wind directions and angles of attack.

Vertical Disp. RMS vs Span Location

600 800 Span Location (ft)

——18 Mode Analysis -- 10 Mode Analysis

-4- Mode 2 Only

--Mode 1 Only

1000 1200

Figure 2: RMS Vertical displacement vs span location: North wind; 0.5% damping

128
Torsional Dlsp. RMS vs Span Location

5.00E-04

4 50E-04

4.00E-04

f 3.50E-04

3.00E-04

J.

2.50E-04 Q

ISI I 2.00E-04

1 150E-04

1 00E-O4

5 00E-05

18 Mode Analysis

10 Mode Analysis

3 Mode Analysis

0 ODE+00 MAiAAtAitAAi

0 203 400 600 Span Location (ft)800 1000 1200

Figure 3: RMS Torsional displacement vs span location: North wind; 0.5% damping

8.00E-KKI

700E+O0

6.00E+00

k

8

4.00E+O0

3.00E+OO

2 00E+O0 •

1.00E+OO

0.0OE-KB

18 Mode Buffeting Analysis

Vertical Displacement Spectra at Mldspan

0.2 0.4

Total RMS Vertical Displacement at Midspan = 4.49 inches

0.6 0.8 Frequency (Hz)

1.2 1.4 16

Figure 4: Vertical displacement spectrum at midpsan: 18 mode analysis; North wind; 0.5% damping

129
3.00E+OO

2.50E+O0

S 2.00E-HX

| 1.50E+00

Q.

£ 1.00E-HM 2

8> 5.00E-01

O.OOE-HX

0.2

18 Mode Buffeting Analysis

Vertical Displacement Spectra at Quarter-Span

Total RMS Vertical Displacement at Quarter Span = 3.07 inches

0.6 0.8 Frequency (Hz)1.2 1.4 1.6

Figure 5: Vertical displacement spectrum at quarter-span: 18 mode analysis; North wind; 0.5% damping

Figure 6. The Fred Hartman Bridge

130
Suu at Midspan - file 200002232

10 -

10

•|Q ' i i i i i i ' ' ' ' ' i i i i i i ;ii

10'3 10'

2 10"1 10° 10

1 102

Frequency (Hz)

Figure 7. Example wind speed spectrum: Measured and predicted.

12

10

E 8

o

6-

Comparison of Measured Midspan RMS Dlsp. (Modes 1&2)

with Analytical Predictions

10 15 20

5-mln average Wind Speed at deck level (m/s)

30

Figure 8. RMS Midspan Displacement in first two modes: Measured and predicted

131
O O.ft 1 l.a 2 2.& 3 3.B 4 4.6 6 O 0.2 O -a OS OS 1 1.2 1.4 1.Q 1.8 2 Time (mln.) Frequency (Hz)

O OS 1 1.S 2 25 3 3.6 4 4.5 6 Time (min .) Figure 9. Time history and power spectral density (PSD) of acceleration for a) deck at midspan (vertical

dir.), b) cable at AS24 (in-plane dir.) and c) deck level wind speed.

132
Strategy on Wind Engineering Research in BRI and NILM

by

Hisashi Okada1

, Yasuo Okuda2, Hitomitsu Kikitsu3 and Masamiki Ohashi4

ABSTRACT

This paper describes current strategy on wind

engineering researches in the Building Research

Institute (BRI) and the National Institute for Land

and Infrastructure Management (NILIM). As for

the strategy on wind engineering researches,

based on performance-based structural design of

buildings, it is significant to evaluate more

reasonable wind loads on buildings by means of

numerical simulations, wind tunnel experiments

and field measurements. The outputs of such

evaluations may be complementary to the new

wind load provision'11 revised in 2000. The

strategy presented here includes various themes of

wind engineering about not only a building

structural aspect but also meteorological or

social/economical aspects.

KEYWORDS: Performance-Based Design

Index of Roughness Terrain

Numerical Simulation

Wind Tunnel Experiment

Strong Wind Damage

High-Rise Building

1. INTRODUCTION

Every year relatively strong and large-scale

typhoon hit Japan, so it is indispensable to design

buildings considering serviceability and safety

limit criteria based on wind engineering. In Japan,

structural designers usually refer several design

regulations and standards on wind engineering,

which are the Building Standard Law of Japan,

Recommendations for Loads on Buildings of

Architecture Institute of Japan (AJJ)[2] and so on. BRI and NILIM have already had important roles

in making out drafts of them. As for the Building

Standard Law of Japan, it was revised in 1998 and

it is a so-called performance-based regulation.

The new wind load provision entrusted by the

Building Standard Law of Japan consists of

several new design concepts based on

Recommendations for Loads on Buildings of AJJ

and other national or international standards such

as IS04354, and the main conceptsm are as

follows:

Setting return period of design wind speed

Introduction of exposure factor and gust

effect factor

Definite separation of loads for design of

structural frames and design of claddings

However, there is still further discussion for

reasonable improvement in the new wind load

provisions.

Under the present circumstances as mentioned

1 Group Leader, Department of Structural Engineering, Building Research Institute, 1 Tachihara, Tsukuba,

Ibaraki, 305-0802, JAPAN

2 Chief Research Engineer, ditto

3 Research Engineer, ditto

4 Researcher, Urban Planning Department, National Institute for Land and Infrastructure Management, 1

Tachihara, Tsukuba, Ibaraki, 305-0802, JAPAN

133
above, it is more important to design buildings

based on performance-based design of wind loads.

BRI and NILIM have set up the strategy on wind

engineering researches for more rational wind load

provisions and some themes of the strategy have

been already started. The outlines of them are

introduced in this paper. We consider that the

following themes will get satisfactory results

under international collaboration with US and

Japan.

2. STRATEGY ON WIND ENGINEERING

RESEARCHES

2.1 Establishment of Design Wind Load Based on

Numerical Index of Roughness Terrain

New wind load provision in the Building Standard

Law of Japan regulates four roughness terrain

categories in Table 1 for the exposure factor as

pointed out in the Introduction. Fig.l shows Er for

each roughness terrain category. For example, it

defines very flat area as Category I and high

density metropolitan area as Category IV ,

respectively. Originally, Recommendations for

Loads on Building of AD as well as other national

or international standards define roughness terrain

categories by descriptive expressions and/or

photographs of typical examples. But the way to

define them was considered to have a possibility of

misunderstanding.

Therefore, in the new wind load provision, a clear

definition was required from the viewpoint of

administrative control of building construction and

it has been developed as follows. The roughness

terrain categories are regulated with considerations

whether the construction site is inside the city

planning area or not and how is the building height

and the distance from seafront or lakefront in Fig.2.

The area of Category I is designated in the area of

Category II , and the area of Category IV is

designated in the area of Category IE, respectively.

It is true that the way to define roughness terrain

categories is distinct administratively, but it is not

always reasonable evaluation from the viewpoint of

wind engineering since the effects of configuration

parameters such as element density 131 are not

evaluated numerically in the provision. Therefore,

we are now planning new research theme

concerning the above problem. Recently the local

height of each building in the roughness terrain has

been able to be measured numerically by means of

remote sensing laser devices on the airplane which

can precisely measure the distances from the

airplane to objects on the ground. Such measured

digital data become available in major cities, the

populations of which are more than 150 thousands

in Japan. They can be classified into two models

which are called Digital Surface Model (DSM) and

Digital Elevation Model (DEM). DSM contains

digital information of absolute heights above sea

level of surface objects such as buildings or trees as shown in Fig.3, while DEM contains digital

information of absolute height above sea level of

the ground level enveloped from DSM. The main

purpose of the theme is proposition of new

numerical index of roughness terrain category

which reflects uneven of the surface of the ground

by means of DSM and DEM.

We have already engaged in measuring wind speed

at some field points. One of them is located at the

rural area in Tsukuba as shown in Fig.4[4]

. The

others are located just near the coast of the Pacific

Ocean in Wakayama and the coast of Tokyo Bay in

Chiba. The roughness terrain category of them may

be different by the possible index, though the

roughness terrain categories of them are the same in

the new wind load provision. Fig.5 shows the power

law index a of the wind profile near the ground

observed at Tsukuba. The roughness terrain

category around Tsukuba also should be IJJ in the

134
new wind load provision, but the results in Fig.5

shows between category II and III . And the

Boundary Layer Wind Tunnel in BRI will introduce

automatic roughness element system in this year. So

we can evaluate validity of the possible index by

both of field data and experimental data. We

consider that the index will be more useful for the

reasonable design wind load.

2.2 Estimation of Urban Disaster by the Strongest

Typhoon

Lots of strong typhoons have hit Japan almost every

year, while the maximum instantaneous wind speed

in history is 105m/s at Guam in 1997 by Typhoon

Paka. From now on we cannot deny the possibility

that such a strong typhoon will hit Japan owing to

the greenhouse effect of the earth. Therefore, it is

important to evaluate the possibility of attack of

such typhoons by means of probabilistic and

numerical simulation. BRI and NILIM have already

started the research in collaboration with Tokyo

Institute of Technology, Kyoto University,

Meteorological Research Institute and Property and

Casualty Insurance Rating Organization of Japan.

The outline of the research is as follows.

2.2.1 Estimation of Possibility about Hitting of the

Strongest Typhoon

In the probabilistic viewpoint, probabilistic data

about the generating area and magnitude of the

typhoon are necessary to collect and then the

probabilistic estimation will be carried out

considering unusual meteorology, the greenhouse

effect and so on. In the numerical viewpoint, on the

other hand, the estimation model of generation of

the strongest typhoon will be developed and

numerical simulation will be carried out. From these

results, it will be possible to discuss how often and

where the strongest typhoon hits, how the structure

of the strongest typhoon is and what meteorological

phenomena including meso-scale cyclone are generated following it.

2.2.2 Estimation of the Generation of Strong Wind

in Urban Area

Recently local digital data of roughness terrain over Japan are going to be available as explained in

section 2.1 and Fig.2. Doppler radar is useful for

measuring wind speed in the sky of urban area'

51

. Moreover, as for the meteorological measuring

network, Atmospheric Environmental Regional

Observation System (AEROS)[6] by Ministry of the

Environment and measuring network by Fire and

Disaster Management Agency as well as the

Automated Meteorological Data Acquisition System

(AMeDAS)m are now provided. We gathered the

surface meteorological data of Typhoon Vicki in

Kansai distinct in 1998 and showed the distributions

of the meteorological elements on the ground'81

. In

this study, we will evaluate how the strong wind

generates in urban area by the numerical simulation

such as Large Eddy Simulation (LES) in Fig.6.

These digital data of roughness terrain and

measured meteorological data are useful as the

boundary condition of this simulation. And then in

order to estimate wind loads on buildings on the

ground, it is important to understand the correlation

of the wind speed above the buildings and the wind

speed around the buildings on the ground. By the

numerical simulation which takes into account such

boundary conditions and the correlation of the wind

speeds in different heights, it is possible to evaluate

tendency of strong wind around buildings and

estimate the degree of the structural damage of

buildings.

2.2.3 Estimation of the Disaster Caused by the

Strongest Typhoon in Urban Area

Based on the results of numerical simulation, it is necessary to estimate not only structural damage of

135
buildings but also social or economical damages in

order to understand the whole damage induced by

the strongest typhoon. Therefore, the main outcome

in this theme is presenting risk management system

against the strongest typhoon in urban area.

2.3 Making out of Database on Buildings Damaged

by Strong Wind

In this study, the main purpose is making out

database of damaged low rise buildings based on

technical reports'

91 on past disaster by strong wind

classifying damaged buildings into several patterns

such as structure, height, shape of roof, roughness

terrain and so on. Since above all low- rise wooden

or lightweight buildings are relatively subject to

attack of the strong typhoon and may end in

collapse as shown in Fig.7, the damages of them

must be discussed at the safety viewpoint as well as

serviceability viewpoint. We consider this study will

be useful for developing precise fragility curve of

risk evaluation system against strong wind and for

limit state design of low-rise wooden or lightweight

buildings.

2.4 Evaluation of Characteristics of Across-Wind

Response of High-Rise Buildings

In order to design structures of high-rise buildings,

wind load is supposed to be more considered rather

than seismic load since across-wind response is

significant under strong wind. In BRI, several wind

tunnel experiments about across-wind response

using tall building dynamic models have been

already carried out and the tendency of the

configuration of building section to influence

across-wind response and aerodynamic damping

effect has been evaluated" 01

.

As next study, it is important to understand how

strong wind velocity surrounding the building

model has an influence on across-wind aerodynamic

responses. In order to understand it, we will

introduce PIV (Particle Image Velocimetry) system

and measure the wind velocities around the building

model by the synchronous measurements of wind

pressures or across-wind deflections. This concept

of a new wind tunnel experiment can be useful for

evaluation of aerodynamic interaction.

Fig. 8 shows an example of a synchronous

measurement of wind velocities around a cubic

model and wind pressures in a wind tunnel

experiment. This measuring system can clarify the

instantaneous complex wind flows around a body

and wind pressure distributions on it, which have

been evaluated only by numerical methods.

2.5 Harmonization of Wind Pressure Coefficient

among Several Codes or Standards

BRI has already held some international workshops

on performance-based design as follows:

• International Workshop on Harmonization of

Performance-based Buildings Structural

Design in Countries surrounding the Pacific

Ocean (Dec. 1997)

[12)

• International Workshop on Performance

Based Building Structural Design(Nov. 2000)

[13]

In these workshops, importance of the

harmonization of design load among countries

was stressed. However, as for wind pressure

coefficient of several codes or standards, there are

differences among them since there are

differences in averaging time of basic wind speeds

and the reference heights of the velocity pressure,

though the basic concepts of them are similar"

41

.

Therefore, from now on we will propose how to

harmonize wind load among codes and standards

in the international meeting like the above

workshops.

136
3. CONCLUSIONS

In this paper outlines of the strategy of wind

engineering researches have been introduced. We

consider that the output of these researches will

contribute to estimating reasonable wind load

affecting low-rise and high-rise buildings on the

basis of performance-based structural design and

also stress that wind engineering area on wind load

which has ever focused on almost only building

structures has necessity to take account of

meteorological or social/economic aspects as well.

REFERENCES:

Pressures on a Building, Summaries of Technical

Papers of Annual Meeting Architectural Institute

of Japan, 2002 (in Japanese).

[12]Proc. of International Workshop on

Harmonization of Performance-based Buildings

Structural Design in Countries surrounding the

Pacific Ocean, BRI, Dec. 1997.

[13]Proc. of International Workshop on

Performance-Based Building Structural Design,

BRI, Nov. 2000(CD-ROM).

[14]R.Yoshie et al., Comparison of peak pressure

coefficients for wind load on cladding in national

and international standards, J. of Wind

Engineering, No.89, 2001.

[I] H.Okada et al., Wind Load Provisions of the

Revised Building Code in Japan, Preprint of the

33

rd UJNR Joint Meeting, 2001

[2] Architectural Institute of Japan,

Recommendations for Loads on Buildings, 1996.

[3] T. Maruyama, On the configuration of roughness

elements in the urban area, J. of Wind Engineering,

No.57, 1993 (in Japanese).

[4] Y.Sasaki et al., Full-scale Measurements on a

Tower-like Structure, Summaries of Technical

Papers of Annual Meeting Architectural Institute

of Japan, 2002 (in Japanese).

[5] http://www.mri-jma.go.jp/Dep/sa/MRI-MSOSRD.html

[6] http://w-soramame.nies.go.jp/

[7] http://tenki.jp/amedas.html

[8] Y.Okuda et al., Weather Situation on Ground

Observed at Fire Stations -Case of Typhoon Vicki,

Proc. of the 32nd UJNR Joint Meeting, 2000

[9] For example, T. Murota et al., Report on Damage

to Buildings Due to Bobara Tornado on December

11, 1990, Kenchiku Kenkyu Shiryo, No.78, 1992

[10]H. Kikitsu et al., Characteristics of Across- Wind

Response of Tall Building with Open Passage,

Proc. of 16

th National Symposium on Wind

Engineering, 2000 (in Japanese).

[II] H. Kikitsu et al., Synchronous Measurements of

Wind Flows around a Building and Wind

137
Tablel Parameters ofEr

Terrain

Category

Zb (m) ZG(m) a

I 5 250 0.10

n 5 350 0.15 i m 5 450 0.20

IV 10 550 0.27

.00

Terrain

Category

0.00 0.50 1.00 1.50 2.00

Er

Fig. J Er (= Vertical Distribution Coefficient for Mean

Wind Speed)

CD

c

3

.a

13m

D5

X

n

m

Example of Digital Surface Model at

Maruno-uchi, Tokyo

F/g.4 Overview ofTower-Like Structure in Tsukuba

a

(a) Outside of

city planning area

Distance from shoreline

orlakefront

(b) Inside of city

planning area

0 6

0 4

0 2

0 0

!

•

Stable

I* • aL±

0 io 0 0

0 1 V

Neutral

i

0

•IV(a=0.27)

n(a=0.15)

I (a =0.10)

10

U (m/s)

RF

Fig. 2 Categories ofExposure Factor

Fig.5 Power Law Index of Wind Profile near the

Ground*1

138
j

Fig.6 Example of Urban Area Simulated by LES with

Local Digital Data of Roughness Terrain,

Maruno-uchi, Tokyo

Fig.7 Example of Building Damaged by Typhoon Bart

(Sep. 1999) The maximum instantaneous wind speed

in Japan was recorded 83.9m/s.

139

Next Generation Wind Tunnels for Simulation of Straight-Line,

Thunderstorm- and Tornado-Like Winds

by

Partha P. Sarkar

1 & Fred L. Haan, Jr.1

ABSTRACT

Researchers at Iowa State University (ISU) are developing next generation wind tunnels for

studying primarily wind effects on structures. The

Wind Simulation and Testing Laboratory in the

Department of Aerospace Engineering and

Engineering Mechanics will house facilities that

simulate straight-line, thunderstorm- and tornadolike winds. This paper describes the motivation for

advancing the state of the art of wind-structure

interaction problems and the work currently

underway at ISU.

A closed-circuit wind tunnel at ISU is currently

being designed to accommodate two test sections

(2.44 m x 1.83 m and 2.44 m x 2.29 m) with

maximum wind speed capabilities of 50 m/s for

aerodynamic testing and 40 m/s for testing in

boundary-layer wind. It will have a gust generator

that is capable of producing gusts up to 125% of

the mean speed and an active lateral-turbulence

generator that is capable of replicating the

turbulence in hurricane winds. In addition, a microburst simulator and a tornado simulator are being designed. While the microburst simulator can produce a stationary/translating downdraft of

1.60 m in diameter and 25 m/s wind speed, the

tornado simulator will produce a stationary or

translating tornado-like vortex of up to 1.22 m in

diameter with a swirl ratio of up to 2.0.

KEYWORDS: atmospheric boundary layer; next

generation wind tunnel; thunderstorm simulation;

tornado simulation; wind loads.

1.0 INTRODUCTION

Building wind tunnels with advanced capabilities

will aid research efforts to understand the complex

fluid structure interaction problems encountered in

wind engineering design. Computer simulations

currently are inadequate for design calculations

because of the complexity of the fluid dynamics

problems involved. Wind tunnels remain an integral

component ofthe design process for wind sensitive

structures.

Whether dealing with the aerodynamics of

buildings, bridges or towers many issues remain to be fully resolved—including the role of nonstationary gust interactions, Reynolds number

effects, and the significance of small-scale

turbulence. Research into these issues is currently

limited by the capabilities of existing wind tunnels.

Building the next generation of such wind tunnels

will contribute to the research infrastructure

necessary to meet the challenges of wind hazards in

this country. Better simulations of atmospheric

flows will enhance our understanding ofthe various

fluid structure interaction phenomena involved and

greatly enhance our ability to develop mitigation measures.

Wind Simulation and Testing Laboratory, Department of Aerospace Engineering and

Engineering Mechanics, Iowa State University, 2271 Howe Hall, Room 1200, Ames, LA 5001 1- 2271 USA

141
Understanding how the construction of wind

tunnels with advanced capabilities constitutes a worthy endeavor requires some background into

the use of wind tunnels in wind engineering and

into the technical problems faced by wind

engineers. This section provides such background

by providing a general introduction to atmospheric

boundary layer (ABL) wind tunnels—the wind

engineer's tool of choice for the past several

decades. Following this introduction, three distinct

types of wind tunnels—for straight-line,

thunderstorm-like and tornado-like winds—will be

described.

2.0 BACKGROUND

2. 1 Boundary-Layer Wind Tunnel

Wind tunnel simulation ofthe earth's atmospheric

boundary layer is a well-established practice.

Numerous researchers have contributed to the set of tools now in use for generating wind tunnel

boundary layers that are several feet deep (for

example, Cermak, 1971; Davenport, 1966; Cook,

1973; Farell and Iyengar, 1999). Conventional

approaches employ a combination of passive

devices such as spires, barrier walls, and floor

roughness to generate boundary layers of the same

scale as the geometric scaling of structural models

placed in them.

It is assumed that atmospheric velocity variations can be adequately modeled by stationary mean and

turbulent flow properties. This assumption means

that despite the fact that hurricanes and gust fronts can have non-stationary characteristics, wind

sensitive structures are tested in stationary flow

environments. Wind tunnel turbulence intensities are matched to site values, and wind tunnel integral

scales are scaled with the geometric scale of the

structural models. While this conventional

approach has served (and still serves) research and

industrial needs for some time, the following two

sections summarize how new tunnel capabilities can answer questions that cannot be addressed with

the current generation of wind tunnels.

2.1.1 Turbulence Effects

The role of turbulence in the relevant fluidstructure interaction problems will influence the

wind tunnel design. While some of the

characteristics ofatmospheric turbulence have been

simulated sufficiently well for some time (for

example, boundary layer velocity profiles, scaling of

turbulence integral scales with model dimensions,

etc.) other turbulence characteristics cannot be

simulated precisely at all or cannot be simulated

without considerable effort. This section briefly

describes the role that turbulence plays in wind

engineering and how new capabilities in wind

tunnels can improve our understanding of these

complex fluid-structure interaction problems.

Civil engineering structures do not, in general, have

aerodynamic performance as their primary design

goal. As a result, most civil engineering structures can be classified as bluff rather than streamlined

bodies. Bluff bodies experience flow separation over significant portions oftheir surface. Bluffbody

aerodynamics differs from aerodynamics of

streamlined bodies in that flow separation and

reattachment play primary roles in pressure

distributions about bodies of interest.

Free stream turbulence can modify the behavior of

shear layers separating from bluff bodies. These

modifications lead to flow structure changes and

pressure distribution changes.

The role of turbulence in the aerodynamics of bluff

bodies has been extensively documented in the

literature (e.g., Gartshore, 1973; Kareem & Cermak,

1979; Hillier & Cherry, 1981; Bearman & Morel,

1983; Nakamura & Ohya, 1984; Kiya & Sasaki,

1985; Saathoff & Melbourne, 1997; and others).

What is clear is that turbulence scales influence

aerodynamic properties (such as rms and peak

pressure coefficients). What is not clear is the extent

of these influences or the precise mechanism of

these influences.

When considering the smallest scales of turbulent

velocity fluctuations, the inertial subrange is a

relevant concept. The "inertial subrange" of a

turbulent flow refers to that range of turbulent eddy

scales between the large inviscid energy-containing

scales and the small viscous diffusion scales. The

size ofthe small, energy dissipating scales decreases

142
with increasing Reynolds number—which

accompanies an increase in the size of the inertial

subrange. The size of the inertial subrange is relevant to bluff body aerodynamics because the

size of the subrange impacts the amount of

turbulent energy residing at small scales. Small

scale content—particularly scales on the order of

the thickness of the separated shear layer—has

been shown by a number of researchers to have a

significant effect on separated shear layer flow

structure (Gartshore, 1973; Tieleman & Akins,

1990).

A significant difference in small-scale turbulence

content can exist between wind tunnel and full- scale flows because wind tunnel Reynolds numbers

can be as much as three orders of magnitude lower

than those of atmospheric flows. To quantify

small-scale content, a "small-scale spectral density

parameter" was originally suggested by Melbourne

(1979) and subsequently used by Tieleman and

Akins ( 1 990). This parameter is essentially a scalespecific turbulence intensity. Tieleman and Akins

reported that wind tunnel simulations with

insufficient small scale content resulted in poorer

comparisons of pressure coefficients between

model and full-scale results.

The above may not fully address all issues relating

to Reynolds number mismatches. Rather, it

illustrates one of the ramifications of failing to match Reynolds numbers in wind tunnel

simulations. Decreasing Reynolds number

disparities between model and prototype flows will

increase our confidence in test results.

Understanding the physics of how flows depend on

Reynolds number will decrease the uncertainty

associated with imperfect turbulence simulation.

Wind tunnels capable ofhigher Reynolds numbers

would enable study of such questions.

2.1.2 Non-Stationary Flow Simulation

In addition to Reynolds number and small-scale

turbulence issues, large-scale turbulent gusts also

constitute an important aspect of wind tunnel

simulation. Passive turbulence generation

techniques (such as the obstacles described

previously) have been shown to produce only a

limited range of possible integral scales

(Bienkiewicz et al., 1983). These scales are often

not large enough to match prototype scales. As a

result, active turbulence generation schemes have

been developed to produce integral scales up to an

order of magnitude larger than those of passive

techniques. These techniques generally involve

grids, flaps, airfoils (and combinations ofthem) that are forced to oscillate (Bienkiewicz et al., 1983;

Kobayashi et al., 1994; Cermak et al, 1995).

While such devices are useful for generating

stationary velocity fluctuations, they have not

generally been used to simulate the non-stationary

gusts that can occur in hurricanes. Anemometry data

from hurricanes has shown that velocity records are non-stationary at times (Schroeder & Smith, 1999).

Thus far, however, no wind tunnel studies have

investigated the impact of such non-stationarity on

aerodynamic pressures on structures. The next

generation of ABL wind tunnel should have the

capability to conduct such tests.

In addition to simulating large-scale gusts, wind

direction changes can be simulated with active

turbulence generation equipment. Wind direction

changes have been observed to significantly affect

pressure distributions on building models in wind

tunnels (Wu et al., 2001a; Wu et al., 2001b). Next

generation ABL wind tunnels will also be used for

furthering research of this type.

2.2 Thunderstorm Winds

Microbursts occur in thunderstorms where the

weight of the precipitation and the cooling due to microphysical processes acts to accelerate the air

downwards. They are characterized by a strong

localized down-flow and an outburst of strong

winds near the surface. Strong outflow winds

develop as the downdraft air is forced to spread

horizontally near the ground level.

Fujita (1985) termed microburst as a small

downdraft having an outburst of damaging winds

with the horizontal extent of the damaging winds

being less than 4 km. This definition has been

modified by radar meteorologists: they require the

peak-to-peak differential Doppler Velocity across

143
the divergent center to be greater than 10 m/s and

the distance between these peaks be less than 4

km.

Thunderstorm winds have significant vertical

velocity components and mean horizontal velocity

distributions different from usual boundary-layer

winds. It is also believed that the gust structure in a downdraft is much better correlated over its width

than in more traditional boundary-layer flow, and

hence will lead to larger overall loading of long

structures. Thunderstorms are responsible for about

1/3 of the extreme gust speeds in the United States

(Thorn, 1969). In recent studies of extreme wind

speeds in the United States, Vickery and Twisdale

(1995) found that, outside ofhurricane regions, up

to 75% of the peak gust wind speeds occurred

during thunderstorms. Selvam and Holmes (1992)

undertook numerical modeling ofthe thunderstorm

downdraft phenomenon, and were able to demonstrate reasonable agreement between a numerical model and limited full-scale data. Later,

Holmes (1999) and Letchford and Illidge (1999)

undertook physical model studies of a jet

impinging on a wall and again found reasonable

agreement between the numerical model, physical

model and full-scale observations of a jet outflow

velocity profile.

2.3 Tornado Winds

Tornadoes are vortices with significant tangential

and vertical velocity components. Therefore, the

flow field in a tornado is much different from the

straight-line boundary-layer wind. Each year

people die and civil infrastructure sustains damage

due to tornados. According to Wind Hazard

Reduction Coalition statistics (for more

information, see www.windhazards.org), each year an average of 800-1000 tornados occur in the U.S.

and cause 80 deaths (on average), 1500 injuries,

and $850 million worth of damage. Although

mostly associated with the region in the central

states often referred to as "tornado alley," tornados

have occurred in all fifty states and also occur in

coastal regions as hurricanes make landfall.

In spite of causing significant losses, tornados have

received little attention from wind engineers.

Statistics show that 90% of all recorded tornados

are rated F2 or less (Bluestein, 1993) on the Fujita

Scale—that is, they involve wind speeds less than

157 mph. It may be economically feasible to design

structures to resist F2 tornados. For cases where

structures cannot be designed to survive, shelters

below or above ground can be designed to protect

people from tornados. It can be argued that certain

essential facilities such as power plants, hospitals,

and airports should be designed for tornados of F3

intensity or higher. Any such design work, however,

requires accurate information about the nature ofthe

wind loads on structures due to tornados.

Determining tornado-induced wind loads is difficult

for two reasons—because quantifying wind velocity

magnitudes in tornados is difficult and because

simulating tornados in a laboratory while measuring

wind pressures on structures is non-trivial and has

not been attempted systematically. Laboratory

simulation of tornados to obtain wind loads on

structures is considered later. With the latest instruments, equipment, and computing facilities, it

is now possible to pursue these goals through

fieldwork and through numerical and laboratory

simulation.

2.3.1 Field Measurements

Beyond the use of storm damage, recent advances in

field measurements have greatly enhanced current

knowledge of tornados and the supercells that

spawn them. For example, the VORTEX project

(which stands for "Verification of the Origins of

Rotation in Tornadoes Experiment") (Rasmussen et

al., 1994) was an effort to use advanced remote

sensing equipment to conduct field measurements of

tornados and tornadic storms.

One particular piece of field equipment that has

proved to be useful is the "Doppler on Wheels"

system (DOW). DOW systems measure wind

velocities directly during a storm from single or multiple truck-based radars (e.g., Wurman, 1998,

Wurman & Gill, 2000). In more recent years two

DOW radars have been deployed in a pattern to allow dual-Doppler analysis. Whereas a single radar can measure the component of the wind along the

line of site of the radar, a dual-Doppler analysis can

determine the 3D characteristics of the flow field.

144
The DOW radars have been upgraded to include a

2.33 m dish, with a 0.9 degree beam width. In a

violent tornado in South Dakota, one of the radars

was deployed within 1 .7 km of the tornado, so that

wind data were collected with resolution of 30 m x 30 m x 38 m (Wurman 1998). These wind

observations from radar are both supporting

existing theories, and in some cases, raising

questions about the previous views of tornado

behavior. For example, Doppler measurements are finding that the decay rate of the wind does not

follow a Rankine curve, as is often assumed

(Wurman & Gill 2000).

2.3.2 Numerical Simulation

Because of prior difficulties in collecting smallscale observations within and near a tornado,

numerical simulations of tornadic flow have been

performed as a tool to improve understanding of

tornado dynamics, small-scale flow characteristics,

and possible genesis mechanisms (e.g., Lewellen et

al., 1997; Lewellen et al. 2000a). Many of the

numerical models developed were patterned after

existing laboratory simulators, and were based

upon simplified forms of the governing equations.

In most of the prior studies conducting numerical

simulation of tornados, the emphasis has been on

processes influencing the general tornadic wind

flow or on potential mechanisms for

tornadogenesis or tornado dissipation. Such issues can be explored with grid spacings larger (roughly

50-100 m) than what is required to examine the

fine-scale details oftornado wind structure near the

ground. Because of the different focus of these

earlier works, little information has been provided

about wind variations near enough to the ground to impact most built structures. For instance,

Lewellen (1993) states that the maximum velocities

will occur below the top ofthe ground boundary

layer (roughly 100 m above surface)—a result that

has great significance for determining tornadoinduced wind loads. Therefore, much more detail

about the near-ground winds is needed.

2.3.3 Laboratory Simulation

Simulating tornados in laboratory environments is not a new concept. Many laboratory simulator

designs have been based on the pioneering work of

Ward (1972). The Ward simulator essentially

consisted of a fan providing updraft at the top of a cylinder above a test area and guide vanes and

rotating screens around the test area to provide

angular momentum to converging flow. Subsequent

efforts—based on the Ward model—at Purdue

University (Church et al., 1979), the University of

Oklahoma (Jischke & Light, 1983) and that of

Davies-Jones (1976) employed various means to improve the similarity between laboratory

simulations and full-scale tornado events.

These laboratory simulations were aimed at greater

understanding ofthe tornado vortex itself. However,

numerical simulation has overtaken physical

simulation as the tool of choice for tornado

studies—both because of cost and because of

versatility. While both laboratory and numerical

simulation efforts have revealed a great deal about

tornado structure (Lewellen, 1993), physical

simulation of tornados for the purpose ofstudying

the tornado may no longer be useful. For the

purpose of quantifying tornadic wind loads on

structures, however, physical simulation is a

necessity.

The fluid mechanical complexity of flow-structure

interaction problems relevant to buildings and

engineered structures require that physical modeling

(usually wind tunnels) be a fundamental design tool. Computers, in most cases, do not have the capacity

for the Reynolds numbers and geometric complexity

involved. This is also true of flow-structure

interactions problems involving tornados.

Chang (1971) and Jischke & Light (1983) both

modified the basic Ward design and added a small

building model with pressure taps. These efforts

found mean surface pressures to be significantly

higher (3-5 times) in swirling, tornado-like vortices

than in straight-line boundary layer flows. This

suggests that when estimating tornado-induced wind

loads on structures, it is not sufficient to use a conventional wind tunnel running with tornado

wind velocities.

3. CURRENT WORK

MS
3.1 Boundary-Layer Wind Tunnel Facility

At this point, much ofthe rough design ofthe wind

tunnel has been completed. Rough sizing and

layout of various components including the test

sections, turning sections, and fan has been

completed (Figure 1). This design incorporates

non-stationary flow capabilities and a high velocity

capacity of 50 m/s (112 mph) along with a large cross section to accommodate realistic models. The

wind tunnel will be of closed-circuit type with the

option of running in an open-circuit mode. It will

have two test sections, one for aerodynamic testing

that is 2.44 m (8 ft) wide by 1.83 m (6 ft) high

followed by a test section to simulate atmospheric

boundary layer wind that is 2.44 m (8 ft) wide by

2.29 m (7.5 ft) high. The maximum speed in the

ABL test section will be 40 m/s (89 mph).

Increased velocity capability will allow larger

Reynolds numbers—with the accompanying

increase in small-scale turbulent spectral content.

A large working cross section will accommodate

both large-scale models and large-scale velocity

structures.

The major component remaining to be designed is the gust front generator that will involve bypass

venting for the wind tunnel. This bypass vent will

connect a portion ofthe tunnel upstream ofthe test section with a portion downstream. Computercontrolled valves will be used to control how much

air is vented through the test section thus

modifying the velocity in the test section.

Preliminary calculations show that variations in

flow velocity of up to 25% ofthe mean wind speed can be obtained.

Other design details include final specifications for

all parts including the fan, the motor, turning

vanes, turbulence screens, honeycomb sections,

etc. The selected fan is of 2.74 m (9ft) diameter

with capability of 220 m

3/s (465,000 cfm) flow

rate driven by a 260 kW (350 hp) AC motor. The

open circuit mode will be achieved by removing

the set of turning vanes at the two successive

corners that follow the test sections. The turning vanes will be moved on rails to the duct that

connects these corners and this duct will be

isolated from rest of the wind tunnel to make the

flow circuit U-shaped. It is also planned to use one

set ofturning vanes as a heat exchanger to minimize

the pressure losses in the tunnel.

3.2 Microburst Simulation

The schematic diagram for a downdraft flow is shown in Figure 2, and the experimental setup for

the microburst simulator is shown in Figure 3. A

honeycomb and two screens were used to reduce the

turbulence of the issuing jet. A 3: 1 area contraction

was used at the nozzle end to make the velocity of

the issuing jet uniform. The diameter (D) of the jet

nozzle was 203 mm (8"). The distance (H) of the

ground plane or impinging platform from the jet

nozzle could be varied from a minimum of 203 mm

(8") to a maximum of 826 mm (32.5"). For the

current work, H/D = 2.89 and H/D = 4.06 were

used. Henceforth, these two heights are termed as H23 and H32, respectively. The H/D ratio for this

study was chosen to be around the median H/D

value of a microburst that varies in between 0.75 to

7.5. Two jet velocities were used, Vo = 9.57 m/s

(31.4 ft/s) and Vo = 16.17 m/s (52.0 ft/s).

Henceforth, these two velocities are termed as V3

1

and V52, respectively. The velocity profiles were measured at distances of 2D and 3D from the center

of the jet and the pressures on the buildings were

also measured at those two locations as well as

directly under the jet (i.e., at 0D). The building

model used was a 25.4 mm (1") cube with 21

pressure taps uniformly spaced along its centerline;

7 taps on each of the front, roof and rear of the

building. Pressures on the ground plane along the

centerline of the jet were measured using 22 taps.

Velocity measurements were done using a DANTEC MiniCTA with a single hot wire. Flow

visualization was achieved using a smoke

generator, and pressure measurements were

conducted using seven Validyne DP-45 transducers

with each having a maximum capacity of ±0.89

inH20.

Smoke flow visualization is shown in Figure 4 for a

stationary jet. The horizontal velocity profiles at 2D

(distance from the center of the jet) are plotted for

two different heights of the jet as shown Figure 6.

146
The profiles are compared with the empirical

profile of Rajaratnam (1976), (Blevins, 1984),

given by

U -0.693[(z-5)/b] 2 . . „

= 1 e

L 1

, z > 8, Eqn.l

U

m

where S is the height z from the ground plane at which the horizontal velocity Uis maximum (Um), and b is the height z where velocity is UJ2 (Figure

5). Also plotted is Wood's (1999) empirical profile

given by

u V

1/6 =

( A'

1.55 1-erf 0.7-

where the variables U, Um, z and b have been

defined earlier and erf( ) is the error function.

There is close agreement between the experimental

and the empirical profiles. They are also compared

with the experimental profile of Letchford &

Illidge (1999).

3.3 Tornado Simulation

Small-scale models of the tornado simulator were

built as design tools for the prototype. These were

built to test concepts that advance the state of the

art of laboratory tornado simulators beyond that

based on the Ward simulator (Ward, 1972).

The first ofthe two simulators consists of a circular

duct (diameter of 356 mm, 14 in.) that is vertically

fixed to a trolley (Figure 7). It is covered at the top.

A motor (0.75 kW or 1 hp, maximum 1750 rpm)

drives a 5 -blade vane located inside the duct

through a pulley and shaft arrangement. The duct

has a clearance of 127 mm (5 in.) from the top

followed by a 102 mm (4 in.)-thick honeycomb

plus screens. The 5-blade vane is 1 52 mm(6 in.) in

length and is located below the honeycomb. There

is a 356 mm (14 in.) gap between the vane and the

exit ofthe duct. There is a deflector mounted at the

duct exit to separate the updraft from the

downdraft that occurs near the periphery of the

duct. A variable speed motor drives the trolley so

that the effect of translation speed on the tornadolike vortex can be studied. The distance between the

ground plane and the bottom of the duct can be

varied up to 584 mm (23 in.). There are two ground

planes that can be separately used in this

experimental setup. The first one has a circular

opening covered by a slotted plate for injecting

smoke or mist from underneath the ground plane to help visualize the flow. Two mist generators were

used to produce mist for visualizing the vortex

(Figure 8). The second ground plane has several

pressure ports connected to a dedicated set of

Validyne transducers to measure pressures on the

ground surface underneath the vortex. The surface

ground pressures were measured underneath the

vortex (Figure 9) and compared with those obtained

using Rankine vortex theory. They are comparable,

but some differences exist. This is expected because

the tornado vortex is somewhat different from a Rankine vortex in the free vortex region near the

ground plane.

For the purpose of exploring different design

configurations of the tornado simulator, a waterbased simulator was also constructed (Figure 10).

Water is often a better visualization medium than

air. The water tank used here is a hexagonal glass

tank. The height of the tank is 0.76 m (30 in.). The

across panel dimension is 0.56 m (22 in.). The basic

design of this simulator is the same as the tornado

simulator mentioned earlier with minor differences.

Due to the small size of this simulator, creating

additional components are relatively inexpensive

compared to the air-based tornado simulator. New

blade assemblies and ducts can be interchanged

with ease. This will aid the design process in terms

of determining the best blade and duct assembly

configuration for producing accurate representations

of tornado vortices.

Visualization is accomplished by introducing glitter

into the water and by using Particle Image

Velocimetry (PIV) whose setup is shown in Figure

10. The velocity vectors were obtained on

horizontal and vertical planes. The tangential

velocity vectors at Z = 25 mm from the ground

plane for the case where a screen (102 mm in

diameter and 102 mm in height) was used without

any surrounding duct at a height of 204 mm from

the ground plane are shown in Figure 1 1

.

147
Table 1 shows several parameters by which

similarity between atmospheric and laboratory

tornado vortices can be quantified. Aspect ratio is defined as the height of the vortex divided by the

radius of the vortex core. Swirl ratio is the ratio of

the vortex circulation compared to the

accompanying flow rate into it. Radial Reynolds

number is calculated using the radial velocity

component of the tornado vortex flow. The

parameters in this table for the ISU prototype

simulator have been estimated based on the design

and should be considered target parameters. The

diameter of the simulated vortex near the ground

for the prototype simulator is expected to be about

1.22 m (4 ft.) for a swirl ratio of 1.0 when the

ground plane is 1 .83 m (6 ft) from the ground. The

diameter of the vortex will increase with reduced

distance of the ground plane from the bottom of

the rotating mechanism. This will allow structural

model scales from 1:100, required for low-rise

buildings, to 1 :500, required for high-rise buildings

and large structures. The surface friction on the

ground plane will be scaled as per the geometric

scale used for each model. The prototype simulator can translate for a distance of 3.05 m (10 ft) at a constant speed of 4.5 m/s (10 mph). The test section will be 6.1 m (20 ft) wide by 1.83 m (6 ft)

high and 9.1 m (30 ft) length.

4.0 SUMMARY

This paper presents the need for the next

generation wind tunnels for simulation of straightline, thunderstorm and tornado-like winds and

describes the current effort that is underway at Iowa State University.

5.0 REFERENCES

Bearman, P.W., Morel, T., "Effect of Free Stream

Turbulence on the Flow Around Bluff Bodies,"

Prog. Aerospace ScL, v. 20 (1983) 97-123.

Bienkiewicz, B., Cermak, J.E., Peterka, J.A.,

Scanlan, R.H., "Active Modeling of Large-Scale

Turbulence." Journal of Wind Engineering and

Industrial Aerodynamics, v. 13 (1983)465-475.

Blevins, R. D, Applied Fluid Dynamics Handbook,

Van Nostrand Reinhold Co, New York (1984).

Bluestein, H. B., "A Review of Tornado

Observations," The Tornado: Its Structure,

Dynamics, Prediction, and Hazards, ed. Church,

Burgess, Doswell, Davies-Jones, American Geo.

Mon. 79 (1993).

Cermak, J.E., "Laboratory simulation of the

atmospheric boundary layer," AIAA Journal, 9, 9,

(1971).

Cermak, J.E., Leighton, S.C., Leffler, R.D., "Windtunnel modeling of the atmospheric surface layer."

Journal of Wind Engineering and Industrial

Aerodynamics, v. 54/55 (1995) 505-513.

Church, C.R., J.T. Snow, G.L. Baker, E.M. Agee,

"Characteristics of tornado-like vortices as a function of swirl ratio: A laboratory investigation,"

Journal ofthe Atmospheric Sciences, v. 36, (1979)

1755-1776.

Chang, C.C., "Tornado Effects on Buildings and

Structures with Laboratory Simulation," Proc. Third

International Conference on Wind Effects on

Buildings and Structures, Tokyo, Japan, (1971)

231-240.

Cook, N.J., "On Simulating the Lower Third of the

Urban Adiabatic Boundary Layer in a Wind

Tunnel." Atmospheric Environment, v. 7 (1973)

691-705.

Davenport, A.G., "The treatment of wind loading

on tall buildings," Proc. Symp. On Tall Buildings,

(1966).

Davies-Jones, R.P., "Laboratory Simulation of

Tornadoes," Proceedings of the Symposium on

Tornadoes, Assessment of Knowledge and

Implications for Man, ed. R.E. Peterson, Texas

Tech University, (1976) 151-174.

Farell, C, Iyengar, A.K.S., "Experiments on the

wind tunnel simulation of atmospheric boundary

layers," Journal ofWind Engineering and Industrial

Aerodynamics, v. 79 (1999) 11-35.

148
Fujita, T. T. The Downburst: Microburst and

Macroburst, The University of Chicago Press,

(1985).

Gartshore, I.S. The effects of free stream

turbulence on the drag of rectangular twodimensionalprisms. Boundary Layer Wind Tunnel

Laboratory 4-73, University of Western Ontario,

(1973).

Hillier, R., Cherry, R.J., "The effect of stream

turbulence on separation bubbles." Journal of

Wind Engineering andIndustrial Aerodynamics, v. 8 (1981)49-58.

Holmes, J.D., "Modelling of extreme thunderstorm

winds for wind loading of structures and risk

assessment," 10th Int. Conf. on Wing Engineering,

Copenhagen, 21-24 June, (1999).

Jischke, M. C, Light, B. D., "Laboratory

Simulation of Tornadic Wind Loads on a Rectangular Model Structure," Proc. Sixth

International Conference on Wind Engineering, v.

1, Australia & New Zealand (1983).

Kiya, M., Sasaki, K., "Structure of large-scale

vortices and unsteady reverse flow in the

reattaching zone of turbulent separation bubble."

Journal ofFluid Mechanics, v. 154 (1985) 463-

491.

Kareem, A., Cermak, J.E., "Wind-tunnel

simulation of wind structure interactions." ISA

Transactions, v. 18, n. 4 (1979) 23-41.

Kobayashi, H., Hatanaka, A., Ueda, T., "Active

simulation of time histories of strong wind gust in a wind tunnel." Journal of Wind Engineering and

Industrial Aerodynamics v. 53 (1994) 315-330.

Letchford, C.W. and Illidge, G. "Turbulence and

topographic effects in simulated thunderstorm

downdrafts by wind tunnel jet", 10

th Int. Conf. on

Wing Engineering, Copenhagen, 21-24 June,

(1999).

Lewellen, W. S., "Tornado Vortex Theory," in

Church et al., eds., The Tornado: Its Structure,

Dynamics, Prediction, and Hazards, Geophysical

Monograph 79, American Geophysical Union,

(1993) 19-39.

Lewellen, W. S., D. C. Lewellen, R. I. Sykes,

"Large-eddy simulation of a tornado's interaction

with the surface,"/. Atmos. Sci., v. 54, (1997) 581-

605.

Lewellen, D. C, W. S. Lewellen, J. Xia, "The

influence of a local swirl ratio on tornado

intensification near the surface," J. Atmos. Sci., v.

57, (2000a) 527-544.

Melbourne, W.H., "Turbulence Effects on

Maximum Surface Pressure - A Mechanism and

Possibility of Reduction." Wind Engineering, Proc.

ofFifth International Conference. Pergamon Press,

(1979).

Nakamura, Y., Ohya, Y., "The effects ofturbulence

on the mean flow past two-dimensional rectangular

cylinders." Journal of Fluid Mechanics, v. 149

(1984)255-273.

Rajaratnam, N., Turbulent Jets, Developments in

Water Science, 5, Elsevier Scientific Publishing

Company, Amsterdam (1976).

Rasmussen, E.N., Straka, J.M., Davies-Jones, R.,

Dowell m, C.A., Carr, F.H. Milts, M.D.,

MacGorman, D.R. "Verification of the origins of

rotation in tornadoes experiment: VORTEX,"

Bulletin ofthe American Meteorological Society, v.

75, (1994) 995-1006.

Saathoff, P.J., Melbourne, W.H., "Effects of freestream turbulence on surface pressure fluctuations

in a separation bubble." Journal of Fluid

Mechanics, v. 337 (1997) 1-24.

Selvam, R.P. and Holmes, J.D. Numerical

simulation of thunderstorm downdrafts", JWEIA,

41-44, pp. 2817-2825,(1992).

Schroeder, J.L., Smith, D.A., "Hurricane Bonnie

wind flow characteristics as determined from

WEMITE." Wind Engineering into the 21

st Century,

Proc. of Tenth International Conference on Wind

149
Engineering, (1999) 329-335.

Thorn, H.C.S. New distributions of extreme wind

speeds in the United States, J. Structures Div.,

ASCE, 94, 1787-1801, (1969)

Tieleman, H.W., Akins, R.E., "Effects of Incident

Turbulence on Pressure Distributions on

Rectangular Prisms." Journal ofWind Engineering

and Industrial Aerodynamics, v. 36, (1990) 579-

588.

Vickery, P. J. and Twisdale, L.A. Analysis of

thunderstorm occurrences and wind speed

statistics, JWEIA, 55, 813-821, (1995).

Ward, N.B. (1972) The Exploration of Certain

Features of Tornado Dynamics using a Laboratory

Model. J. ofthe Atmospheric Sciences, 29, 1 194-

1204.

Wood, G. S and Kwok, K. C. S. "Physical and

Numerical Modeling of Thunderstorm

Downbursts", Wind Engineering into the 21

st

Century, Proc. of the 10th Intl. Conf. on Wind

Eng., eds. A. Larsen et al., Denmark, June (1999),

pp. 1919-1924.

Wu, F., Sarkar, P.P., Mehta, K.C., "Full-scale

study of conical vortices and roof corner

pressures." Wind & Structures, v. 4, n. 2 (2001)

131-146.

Wu, F., Sarkar, P.P., Mehta, K.C., Zhao, Z.

"Influence of incident wind turbulence on pressure

fluctuations near flat-roof corners." Journal of

Wind Engineering andIndustrial Aerodynamics, v.

89, (2001)403-420.

Wurman, J. "Preliminary results from the

ROTATE-98 tornado study," Preprints, 19th Conf. on Severe Local Storms, Minneapolis, MN, 14-18

September, (1998) 120-123.

Wurman, J., Gill, S. "Finescale radar observations

of the Dimmitt, Texas (2 June 1995), tornado,"

Mon. Wea. Rev., v. 128, (2000) 2135-2164.

150
TABLE 1. LABORATORY TORNADO SIMULATOR PARAMETERS

Swirl Ratio Aspect Ratio

Radial

Reynolds No.

Atmospheric Range 0.050-2.0 0.2-3.0 Ie9-lell

Dallas Tornado 0.8 3.0 2e9

Purdue Simulator 0.1-1.0 0.3-3.0 4.1e3-1.2e5

Oklahoma Simulator 0.0-0.8 1.2

ISU Simulator 0.1-2.0 0.5-3.0 up to le6

Figure 1 Three-dimensional rendering of the design for the atmospheric boundary layer wind tunnel.

151
Figure 2. Schematic view of a typical downdraft flow showing the developed boundarylayer profile of horizontal flow field as compared to a straight-line boundary-layer wind

profile

Figure 3. Experimental setup for the model microburst simulator

152
Figure 4. Microburst flow visualization with smoke

Figure 5. Boundary-layer horizontal velocity profile in an impinging jet

153
HORIZONTAL VELOCITY PROFILE AT 2D, H32,

0.0 0.5 1.0 1.5

NORM. VELOCITY

Figure 6. Comparison of normalized velocity profiles at 2D distance from the center

of the microburst (D diameter). EXP. ISU experiment, TH. Blevin's (1984) Profile

Figure 7. View of the model microburst/tornado simulator (1 :5 scale) at ISU

154
Figure 8. Tornado flow visualization with mist (fine water particles in air)

Ground Pressure for Tornado (H = 152 mm), Radius of Core, Rc

= 41 mm, Peak Pressure = - 5.6 mm Water

Distance from Center (mm)

Figure 9. Normalized pressure distribution as measured underneath the tornado-like vortex on

the ground plane compared with Rankine vortex

155
Figure 1 0. Experimetal setup for PIV measurement for flow visualization and

measurement in water simulator

-120 -80 -40 0 40 80 120 160 [»»)

Figure 1 1 . Tangential velocity vectors in the water simulator at z = 25 mm from ground plane. A

screen (102 mm D and 102 mm H) without duct at 203 mm from ground plane was used to generate the vortex.

156
SESSION 4

DAMS

Seismic Performance Evaluation of Intake Towers

by

Richard C. Dove1 and Enrique E. Matheu

ABSTRACT

In the event of an earthquake, it is vitally

important that the catastrophic failure of a dam

and subsequent sudden release of the reservoir

be prevented. An important part of the

prevention of such a failure is maintaining the

ability to control the release of water after the

earthquake. For most earthen dams, and some

concrete dams, the release of water is controlled

through a reinforced concrete intake tower. The

functional survival of such towers has been the

main concern of a multi-year research effort

sponsored by the U.S. Army Corps of Engineers

(USACE). Most intake towers in the current

USACE inventory are lightly reinforced. The

functional survival of such lightly reinforced

structures is thus the main concern of this

research effort. The ultimate objective of this

research work is the development of analysis

procedures for seismic evaluation of these

structures. This paper presents some of the

results of this effort.

KEYWORDS: Reinforced concrete, ductility,

intake towers, seismic evaluation, displacementbased approach, shaking table tests.

1.0 INTRODUCTION

Earthquake engineering research efforts in the area of reinforced concrete intake towers are currently focused on understanding the nonlinear

response of lightly reinforced intake towers. The

ultimate objective is the evaluation and/or

development of simplified analysis procedures

for the seismic evaluation of these structures. The work presented in this paper is part of a

larger research effort that began with a statistical

analysis of the USACE inventory of existing

intake towers. This tower inventory analysis

quantified the distribution and variation of the

structural characteristics of the towers as relating

to their earthquake location hazard (Dove,

1996). The information collected was used in

planning the second phase of this research

program, referred to as the Intake Tower

Substructure (ITS) experimentation series

(Figure 1), which was conducted during 1996

and 1997 (Dove, 1998) at the Geotechnical and

Structures Laboratory (Vicksburg, Mississippi).

Figure 1. Typical 1/8-scale static experiment.

The results from this experimental effort, which

included monotonic and cyclic loading tests of

1/8-scale models, showed that substantial

ductility is available (Figure 2). The objectives

of these experiments were not only to observe

the response of reduced scale models of typical

intake towers and quantify the ductility

available, but also to use the information

generated for the development of approximate

and/or simplified evaluation procedures for

existing intake towers.

1 Geotechnical and Structures Laboratory, U.S. Army Engineer Research and Development Center, Vicksburg, MS

39180 (USA).

2 Department of Civil and Environmental Engineering, Louisiana State University, Baton Rouge, LA 70803 (USA).

159
•ao

45000

30000

15000

0

-15000

-30000

-45000

rff.g^- _

!'

1

-1 6 -1.2 -0.8 -0.4 0 0.4 0.8 1.2 1.6 2.0 Top Deflection

Figure 2. Typical load-deflection curve

exhibiting substantial ductility.

Based on the results of the ITS experiments, it was concluded that additional information was

required for the application of the simplified

analysis procedures under development.

Specifically, a method was needed to properly

estimate the ultimate deflection capacity of

existing intake towers. During 1999 and 2000,

fourteen 1/2-scale experiments were conducted

to provide a statistically significant basis for the

development of an empirical estimation of the

parameters needed (Dove, 2000). As a result of

these tests, a modified displacement-based

analysis procedure was generated incorporating

the available experimental information. This

analysis procedure is discussed in the next

section. More recent research efforts have

mainly focused on the validation and calibration

of the proposed analysis procedure and its underlying assumptions by means of an

extensive series of shaking table tests. The first series of tests were recently performed at the

facilities of the Construction Engineering

Research Laboratory (Urbana, Illinois). The

corresponding results and preliminary

conclusions are discussed in following sections.

2.0 DISPLACEMENT-BASED ANALYSIS

The nonlinear response and ductility of lightly

reinforced intake towers has been the focus of

recent analytical and experimental efforts. It has

been shown that lightly reinforced intake towers can exhibit ductility but with a very localized

failure. When a lightly reinforced intake tower is excited by a seismic event, a single crack forms

at the base of the tower or at the location of a major stiffness change. Experimentation has

shown that ultimate failure is dependent on the

response of the rebar within the crack (Dove,

2000). An analysis technique has been

developed that reflects this localized failure

mode and includes explicit consideration of the

earthquake-induced displacements of a structure.

It also attempts to account for the shift of the

fundamental frequencies with formation of

plastic regions in the structure. The proposed

analysis procedure will be presented by applying

it to a representative rectangular intake tower.

The structure analyzed has been used as an example problem in past and current USACE

guidance documents (U.S. Army Corps of

Engineers, 2002). The analysis presented here is essentially a modification of the prior analysis

presented in this reference. The structure is a generic tapering rectangular tower about 61 m

tall, 14.6 m by 11.6 m wide with a wall

thickness 1.83 m at the base, as shown in Figure

3. Primary reinforcement consisted of #11 bars

at 30.48 cm on center.

7.92m-

(29.0')

7.62m-

(250')

.61m -

(2.0')

NORMAL —

\

POOL \

EL 1016.81m\

(3336')

,91m-

(3.0')

1.22m-

(4.0')

1.52m-

(5 0')

z 1.83m

1 (6.0')

• j

• q

• f

L

(3400)

EL 1024.13m

(3360')

(3320')

EL 999.74m

(3280')

i LUMPED MASS

LOCATIONS

.b-H

i • 0

U-i,

I 1 (5. T

I?DO id

UPSTREAM ELEVATION

(3240')

68m

5')

(44.0')

12.91m

(40.0')

(3200')

SIDE ELEVATION

Figure 3. Layout of example intake tower.

160
The assumed analytical model consists of a simple cantilever beam attached to a rotational

spring. The spring approximates the response of

the cracked region. The beam models the

response of the uncracked tower above the base.

The definition of the rotational spring stiffness

requires the calculation of the moment-curvature

(M-(|)) relationship. Based on this relationship, it

is then possible to calculate the corresponding

moment-rotation relationship (M-0), which is obtained by multiplying the curvature by an assumed plastic hinge length. The M-9

relationship represents the stiffness of the

rotational spring.

The M-6 relationship is often strongly bilinear,

and therefore a simplification is required in

order to conduct a response spectrum analysis.

The M-0 relationship is linearized such that it encloses the same area for the same maximum

rotation. This approximation allows the

calculation of the expected deflection under the

given earthquake loads. Given the linear spring

stiffness, the element properties, and any added

mass due to water, a response spectrum analysis can be readily conducted. The maximum

deflection calculated represents the deflection

demand of the tower under the input earthquake.

Deflection Rotation

Figure 4. Deflection capacity.

In order to complete the analysis, it is necessary

to determine the ultimate deflection capacity 6 U (Figure 4), which is calculated as follows:

5„=^+ 5 Ctt (1)

where § E

is the ultimate elastic curvature at the

base of the tower, also known as the cracking

curvature; Q P represents the plastic rotation at

failure; l w denotes the depth of the section, L is

the height of the intake tower; and b cu represents

the deflection contribution caused by the base

crack.

This model assumes that the ultimate lateral

deflection consists of the sum of two parts. The

first part is the elastic response of the body of

the intake tower above the cracked section. The

elastic curvature § E can be computed as M

<y*"E *£=

i?7 <9

where M denotes the yielding moment, E the

elastic modulus, and I

g

the uncracked moment of

inertia. The second part is a rigid body rotation

of the tower as the crack opens at the base of the

elastic section, and the tower rotates about the

neutral axis of the cracked section. It is conservative to assume that the neutral axis is coincident with the edge of the tower. Hence,

the lateral rigid-body deflection at the top of the

tower varies directly with the crack width, and

its maximum value is as a ratio of the tower

height and the tower width, times the ultimate

crack width. Therefore,

5 CU =QP L=^L (3)

The principal unknown in the above equation is cu , i. e., the ultimate crack width at failure, which

is a function of the ultimate strain of the rebar

and the strain penetration. Experiments have

indicated that for a single crack response, the

crack widths are largely predicted by the

ultimate strain capacity of the rebar and rebar

diameter. An empirical equation for cu was

generated as follows:

cu =0.12 + 2.47s u +0.79db (4)

where z u is the ultimate strain at failure of the

rebar as measured over the standard gage length

(20.3 cm or 8 inch) and db is the diameter of the

reinforcing bar in cm.

161
Application of the proposed displacement-based

procedure to the design example begins with the

calculation of the M-(() relationship for the

bottom section of the tower in two directions

(weak and strong axis). This follows the

assumption that the failure mechanism of the

tower will be the formation of a single crack at the base of the bottom section. The vertical dead

load of the tower self weight was included.

Assuming an 18 percent ultimate strain, the

ultimate crack width calculated from the above

empirical equation to be 1.40 cm. The strain

penetration length (Ls ) can be calculated from

the following equation:

This gives a strain penetration length of 7.52 cm.

Multiplying the M-(j) diagram by this strain

penetration length gives the corresponding M-0

relationship. Proceeding as described above, the

equivalent rotational spring stiffness is calculated such that the total area under the

equivalent rotational spring is the same as that

for the corresponding weak and strong axis M-0

relationships. The resulting equivalent spring

constants are given by 2.107E+12 N-m/rad and

3.561E+12 N-m/rad for the weak and strong

axes, respectively. Based on these properties,

and assuming a finite-element discretization of

the structure using standard beam elements, a

response spectrum analysis was conducted using

the MDE response spectrum for 5% damping

indicated in Figure 5. The results of this analysis

indicate a top deflection of 9.6 cm for rotation

about the strong axis and 10.1 cm for rotation

about the weak axis.

Finally, it is necessary to calculate the ultimate

deflection capacity 5„ . Given the ultimate crack

width and section width at the base of the model,

the ultimate base rotation Q P can be calculated.

The elastic curvature <t> £

at the base of the intake

tower (also known as the cracking curvature) is determined from the M-<j) relationship for the

section. Based on these parameters, the

deflection capacity of the tower is calculated as 9.6 cm for rotation about the strong axis and

12.6 cm about the weak axis. Therefore, the

tower passes the analysis.

_ 0.6

0

11 \* 0-62g

V—MDE

0J31g v Sa —

T

1 \< \

O30g

' 0.l5g Sa T

I I

'

' ' -i

0 0.5 1 1.5 2 2.5 3

Period - T (seconds)

Figure 5. Standard spectra used in analysis.

3.0 SHAKING TABLE EXPERIMENTS

As part of a continuing effort to validate the

displacement-based analytical procedure, a

series of shaking table experiments was

performed. The experiments were conducted on

a 1/8-scale model of a typical intake tower, of

the same design as used for the ITS3 cyclic

loading tests. The objective of these tests was to compare the failure mode under dynamic

conditions with the failure mechanism

previously observed under monotonic and cyclic

loading. These tests also served the purpose of

providing additional data for the evaluation and

validation of current analytical models.

Figure 6. Intake tower model on top of shaking

table at the Construction Engineering Research

Laboratory (Urbana, Illinois).

162
The tests were completed by early July 2001.

The intake tower model (Figure 6) was tested

laterally with uniaxial sinusoidal support

motions, near the tower's natural frequency in

the short direction of the tower. Test levels were

increased until failure occurred. Pre- and postfailure responses of the model as well as the

corresponding failure mechanism were

systematically documented. The extensive

results collected from this testing program will

facilitate a direct comparison between the

measured responses and the behavior predicted

by numerical models and previous static cyclic

tests.

The model was extensively instrumented with

accelerometers attached to the base beam and

the external faces, linear variable differential

transducers (LVDT) installed between the base

beam and tower, and several deflection gages. In

addition, the vertical reinforcing steel was

instrumented with strain gages. Progression of

damage was extensively documented with

digital photographs and video. Two digital

camcorders were used: one recorded the overall

model response from the east-north-east side,

whereas the other camcorder zoomed in on the

cold joint at the bottom of the model on the west

face of the model (Figure 7).

The intake tower model was 3.05 m tall with a hollow rectangular cross-section that was 1 .32 m

wide in the east-west direction and 1 .02 m wide

in the north-south direction. The walls were 0.14

m thick, constructed of normal strength concrete

with scaled aggregate and reinforcing steel. The

weight of the tower was calculated as 4.29

metric tons. Figure 7 shows a schematic drawing

of the intake tower model, which was placed on

top of a heavily reinforced base beam. The base

beam was square, with a width of 2.74 m and a thickness of 0.46 m.

E

in

Figure 7. Schematic view indicating intake

tower model dimensions.

Table 1. Scaling relationships

Variable

Time

Frequency

Force

Stress

Strain

Displacement

Velocity

Acceleration

Scaling factors

XL , XE , X

XF - XEXL

X, =

x;! =

x*

hL ,XE -Xp -\

Xf — X}

x f =

XF - XL

Xe = l

Xa =l

Table 1 displays the scaling relationships that

govern this type of problem, in terms of 3 main

scaling factors: X L (geometry), X E (modulus of

elasticity), and X p (density). The model

geometry was 1/8 of the prototype, that is,

XL = 8 . Standard strength concrete was used for

the model, and therefore XE = X

p = 1 . The third

column of the table contains the resulting scaling

relationships for this case, which is characterized

by an acceleration scaling factor that is the

reciprocal of the geometric scaling factor.

Therefore, inertial and gravity effects for the

163
model should be increased "k^ = 8 times with

respect to prototype conditions. The scaling of

inertial effects was achieved by increasing the

magnitude of the imposed base accelerations by

a factor of 8.

Cold Joint

r 1iini

\ 6389 I

Figure 8. Schematic view depicting

modeling of gravity load effects.

The scaling of gravity effects was achieved by

means of 80 loops of 2.54cm diameter elastic

cord connected between the top and the base of

the tower model (Figure 8). The extremely flat load-deflection characteristics of these elastic

cords allowed them to be used in such a way that

the corresponding vertical loads did not vary

significantly with the rocking response of the

tower with respect to the base. The cords were

designed to provide an additional vertical load

equal to 30 metric tons. Including this load, the

resulting vertical compressive stresses at the

base of the model due to gravity effects were

575 kPa.

The model was initially tested with low-level

random motions to measure natural frequencies,

mode shapes and damping. These tests were

conducted in the X (north-south) and Y (eastwest) directions. The amplitude of the input

motions were initially very low (0.02 g). The

random motion tests were then repeated at increasing amplitudes until clear modal

information was acquired in both horizontal

directions.

The model was initially excited using sinusoidal

base motions in the X direction at a frequency of

28 Hz (3.5 Hz, prototype scale). The imposed

base motions ramped up to full amplitude in 0.5

seconds; held a constant amplitude for 2.0

seconds; and ramped down in 0.5 seconds, for a

total test duration of 3.0 seconds. For each test run, the amplitude of the sinusoidal support

motion was gradually increased with respect to the previous one while keeping the excitation

frequency constant. Previous static tests performed on similar 1/8-scale models indicated

that the base of the model cracked at a value of

top lateral deflection between 1.3 and 1.8 mm,

with applied lateral loads between 130 and 170

kN.

The model failed by cracking at the cold joint at the base. Once it cracked across the entire cold

joint surface, the model softened significantly.

The change in the fundamental frequency was a very effective damage indicator. Damage

evolution was carefully monitored by the LVDT

measurements along the base of the model.

12 "I ;

1

8

-

c* 4

-

-12-1 i r 1 1 1

1 0.00 0.50 1.00 1.50 2.00 2.50 3.00 Time [sec]

Figure 9. Base acceleration, A6x (run 18).

12 "j „ 8

-

-12-1 1 1 1 1 1

1 0.00 0.50 1.00 1.50 2.00 2.50 3.00 Time [sec]

Figure 10. Top acceleration, A14x (run 18).

Figures 9 and 10 show acceleration responses

measured during test run 18, corresponding to an

excitation level of 2.04 g (0.255 g, prototype

164
scale). These responses correspond to sensors A6x and A14x, located at the bottom and top of

the north face of the model, respectively. Figure

11 shows the time history measured by one of

the deflection gages installed at the top of the

model (D5x). The amplitude of oscillation is

about 2 mm.

8 T

6

- ? 4-

-8 i " 1 1 1 1 1

0.00 0.50 1.00 1.50 2.00 2.50 3.00 Time [sec]

Figure 11. Top deflection, D5x (run 18).

After significant damage was identified, the

model was again tested with random motions to

determine updated modal properties associated

with the damaged condition. Cracking of the

model across the entire cold joint surface

softened the structure, as evidenced by the shift

in the fundamental vibration frequency. Figure

12 shows the transfer functions between the top

model acceleration (A14x) and the base beam

acceleration (Alx) corresponding to two

different series of random motion tests. As

shown in the figure, the fundamental frequency

decreases from 29 Hz to 24 Hz.

/A \ Test 1 // Vv

Test 2 I

0 25 50 75 100 125 Frequency [Hz]

Figure 12. Transfer functions A14x/Alx for

different series of random excitation tests.

Additional sinusoidal tests were conducted with

an excitation frequency of 22 Hz (2.75 Hz,

prototype scale), gradually increasing the

amplitude of the motions as in the previous case. Figures 13 and 14 show acceleration responses

measured at the top and bottom of the model

during test run 37, corresponding to an

excitation level of 6.00 g (0.75 g, prototype

scale).

12 T — 1

-12 -I 1 . 1 1 11 0.00 0.50 1.00 1.50 2.00 2.50 3.00 Time [sec]

Figure 13. Base acceleration, A6x (run 37).

12"1 ;

-12 -I 1 r- .

i

1 1 0.00 0.50 1.00 1.50 2.00 2.50 3.00 Time [sec]

Figure 14. Top acceleration, A14x (run 37).

As seen in these figures, the behavior of the

model at this stage exhibits different response

characteristics that are induced by the significant

damage and rotation at the base. Similar

conclusions can be drawn by considering the top

deflection of the model, shown in Figure 15.

0.00 0.50 1.00 1.50 2.00 2.50 3.00 Time [sec]

Figure 15. Top deflection, D5x (run 37).

165
Extensive quantitative information regarding the

behavior of the failure surface (cold joint at the

base of the model) was gathered by the vertical

LVDT gages at that location. Figure 16 shows

the time history for the sensor L2Z, illustrating

the opening and closing of the single base crack.

1.20

-0.40 -I 1 1 1 1 1

0.00 0.50 1.00 1.50 2.00 2.50 3.00 Time [sec]

Figure 16. Base LVDT, L2z (run 37).

4.0 CONCLUSIONS

These experiments demonstrated that the failure

mode under dynamic conditions was very

similar to the failure mechanism previously

observed under monotonic and cyclic loading.

The same single crack response with significant

ductility was witnessed in both the static and

dynamic experiments. As expected, the natural

frequency of the model decreased substantially

after cracking. The experiment validates the

calculation of the deflection capacity based on

the results of static experimentation. Future

dynamic experiments will model the response of

a tower to an individual earthquake. This will

further validate the complete displacementbased analysis process.

5.0 ACKNOWLEDGEMENTS

This work was sponsored by the Direct-Allotted

Civil Works Earthquake Engineering Research

Program of the U.S. Army Engineer Research

and Development Center (ERDC) in Vicksburg,

Mississippi. The authors appreciate the

cooperation of the authorities at the ERDC,

which permitted the presentation and publication

of this study. Permission to publish this paper

was granted by the Chief of Engineers, U.S.

Army Corps of Engineers.

6.0 REFERENCES

Dove, Richard C. (1996). "Structural Parameter

Analysis of U.S. Army Corps of Engineers

Existing Intake Tower Inventory," Technical

Report SL-96-1, U.S. Army Engineer Waterways

Experiment Station, Vicksburg, MS.

Dove, Richard C. (1998). "Performance of

Lightly Reinforced Concrete Intake Towers under

Selected Loadings," Technical Report SL-98-1,

U.S. Army Engineer Waterways Experiment

Station, Vicksburg, MS.

Dove, Richard C. (2000). "Performance of

Lightly Reinforced Concrete Intake Towers under

Selected Loadings," Technical Report ERDC/SL

TR-00-6, U.S. Army Engineer Research and

Development Center, Vicksburg, MS.

Headquarters, U.S. Army Corps of Engineers

(2002). "Structural Design of Spillways and

Outlet Works," Engineer Manual EM 1110-2-

2400, Washington, DC (in publication).

166
Present State of Measurement of Earthquake Motion and

Nation-wide Networking of Seismographs at Dam Sites in Japan

by

Tadahiko Sakakomo1

, Syuji Takasu2

, Hitoshi Yoshida3

, Yoshikazu Yamaguchi4

, Takashi Sasaki

5 and Kenji Inagaki

6

ABSTRACT

The SMAC-type strong motion recorder was

installed in the Sarutani Dam in the Kinki Region,

Japan in 1957, and strong motion observation on dam structures under the jurisdiction of the MLIT,

the Ministry of Land, Infrastructure, and Transport

(formally the Ministry of Construction) had been

started. Since the Niigata Earthquake in 1964, the

necessity of strong motion observation has been

widely recognized and the Ministry has been

positively promoted the installation of seismometers

in dams.

Nowadays, the 413 dams under the jurisdiction of the

Ministry have seismometers and the high-density

observation of earthquake records can be conducted.

Furthermore, the network of observation of

earthquake records at dam sites has been speeded, the

236 seismometers at 50 dam sites have been

connected to the Public Works Research Institute

(PWRl), presently the National Institute for Land and

Infrastructure Management (NILIM).

This report describes the present state of earthquake

data measurement at dams under jurisdiction of

MLIT, the characteristics of collected data, and the

outline of the dam-site seismograph network that is

now being constructed.

KEYWORDS: Strong motion records, Dam,

Seismometer, Network of observation of earthquake

record

1. INTRODUCTION

Many large earthquakes struck Japan and caused

severe damage on important infrastructures and

private properties. Fortunately, no dams in Japan has

suffered damage that threatened its safety, but

clarifying and securing the earthquake safety of dams

is one of the most important missions that must be

faced in densely populated Japan. To respond to a

strong demand for the advancement of evaluating

methods of the earthquake safety of new and existing

dams, it is important to study the characteristics of

seismic motion at rock foundation and the dynamic

behavior of actual dams based on observed records

measured at various dam sites. Installing

seismographs at dams have been speeded since the

1995 Hyogo-ken Nanbu Earthquake (the Kobe

Earthquake) and all dams have seismographs under

jurisdiction of the MLIT now.

1 Chief Executive, Public Works Research Institute (PWPJ), 1-6 Minami-hara, Tsukuba-shi, 305-8516, JAPAN

2 Group Leader, Hydraulic Research Group (HRG), PWRI

3 Research Coordinator for Watershed Management, River Department, National Institute for Land and

Infrastructure Management (NILIM), Ministry of Land, Infrastructure, and Transport (MLIT), 1 Asahi,

Tsukuba-shi, 305-0804, JAPAN

4 Head, Dam Structures Research Team, HRQ PWRI

5 Senior Researcher, Dam Structures Research Team, HRQ PWRI

6 Researcher, Water Management and Dam Division, River Department, NILIM, MLIT

167
2. PRESENT STATE OF MEASUREMENT

OF EARTHQUAKE MOTION AT DAMS

IN JAPAN

2.1 Background

The measurement of earthquake motion at dams

began in 1957 when the Ministry of Construction

(presently the MLIT) installed a SMAC-type strong

motion seismograph at the Sarutani Dam to obtain

data vital for dam safety management and to advance

seismic resistant design. The installation of strong

motion seismographs was promoted in response to a

growing awareness of the need for strong motion

observations following the Niigata Earthquake in

1964.

At the Kobe Earthquake in 1995, a lot of several

strong motion records at dam sites were observed,

but some dams near the earthquake faults didn't have

seismometers, so the strong motion records at dam

sites near the earthquake faults were not got

unfortunately. The Ministry, therefore, decided to

place the seismometers in all dam sites under its

jurisdiction, and to replace the old-fashioned analog

seismometers to the new digital seismometers. By

the end of 2000, all dam got the seismometers.

2.2 State of installation of strong motion

seismographs

By the time of the Hyogo-ken Nanbu Earthquake,

443 strong motion seismographs were installed at

140 dams (MLIT's dams: 48, Water Resources

Development Public Corporation's dams: 10,

Subsidized prefectural dams: 82) , but 1,152 have

now been installed at 413 dams (MLJT's dams: 78,

Water Resources Development Public Corporation's

dams: 20, Subsidized prefectural dams: 315) . Figure

1 shows the locations of the dams where the

installation of strong motion seismographs has been

completed, and Figure 2 shows changes in the

number of strong motion seismographs in operation.

2.3 Standard installation method of seismographs at

dams

Strong earthquake motion observations at dams play an important role in dam safety management and

studies of the response of the dam to strong motion in

bedrock. Therefore, in principle, installation of

seismographs is done at the base and at the crest of

each dam in the maximum cross section. But

because differing responses appear at bridges and

pier structures on dam crest, installing seismographs

at these locations is avoided generally. Table 1 shows

a breakdown of strong motion seismograph

installation locations. Dam base mainly means the

gallery near the base of concrete dams or the gallery

beneath the impervious zone of rockfill dams. Other

locations include rim grouting tunnels, water intake

systems, and management offices. Data observed at

foundations under the dam include reflected waves

from on the dam body. Therefore, installation of

seismographs on the downstream free field is recommended !)

. The components measured by a seismograph are set

of course as 2 horizontal components and 1 vertical

component. Vibration of dam body occurs most

easily at right angles to the dam axis because of the

vibration characteristics of dams. Therefore, the

directions of the horizontal components are, in

principle, right angles to the dam axis and the dam

axis direction.

Earthquake observations are normally performed in

trigger mode; the seismographs are all linked and

begin recording simultaneously when a certain

standard value is exceeded. In general, trigger

seismograph is located at dam base.

Strong motion seismographs that are now installed at dam sites include analog type seismograph and

digital type seismographs. Figure 3 shows the state of

installation of analog type and digital type

seismographs. Initially, most were analog type

equipped with a moving pendulum to record

earthquake motion on paper, but most of those now

in use are digital type that converts earthquake

motion to electrical output that is then digitized by an AD converter and recorded on an IC card etc. In the

168
case of new installation, only digital type

seismographs are now selected.

3. CHARACTERISTICS OF ACCELERATION

AT ROCK FOUNDATION OF DAM SITE

Two reports about the collection of observed

acceleration data at dam sites by the PWRI were

published in January, 2001 . The first report

21 is about

the information of the locations and the specifications

of seismographs in each dam at the time of 2000.

The second report

3) summarizes the acceleration data

and spectrum and information of earthquakes. The

number of records dealt with in this report is 2,960

collected in from 1966 to 2000.

In this paper, the characteristics of horizontal

acceleration records in dam foundations have been

examined based on records of observed earthquake

data. Dams under the jurisdiction of the MLIT were

all constructed on rock foundations. The S-wave

velocities of dam foundations are therefore, expected

to be at least 1 ,000 m /sec.

The 143 sets of, 286 in total of horizontal

acceleration records were used in this study. These

data were collected during earthquakes with a

magnitude of 5 or more and observed at an epicentral

distance of 200 km or less. Figure 4 shows the

distribution of the earthquake epicenters and Figure 5

shows the relationships of the magnitude, hypocenter

depth, and peak acceleration with the epicentral

distance. The distribution of observed results is

relatively dispersive.

Figure 6 shows the relationship of the epicentral

distance with the peak acceleration. The data are shown with the presumption values by the

attenuation relationship equations by Fukushima et

al.^ and An-naka et dlf These two attenuation

relationship equations are based on the data mainly

collected on hard soil foundation. The data of dam

sites seem to be small as compared with the

presumption values.

Figure 7 shows acceleration response spectra. And,

Figure 8 shows acceleration response magnification

spectra.

Figure 9 shows acceleration response spectra for

different epicentral distances. Figure 10 shows

acceleration response magnification spectra obtained

by dividing acceleration response spectra by the peak

acceleration values. In order to investigate the

characteristics of the relatively strong motions, the

data shown in these figures are reduced data

observed earthquakes of a magnitude 6 or more and

within 100 km of the epicentral distance. The

number of extracted data is 122. The peak value of

response magnification spectrum is about 2 for all

earthquake magnitude and epicentral distance, and

the peak values mostly appear within the periods of

0.1 -0.2s. The response magnification spectrum

decreases rapidly for the period over 1 sec, and the

larger earthquake magnitude is, the smaller the

decreasing rate of the response magnification

spectrum according to the period is.

4. DAM SEISMOGRAPH NETWORK

Because the observed data are very valuable and

indispensable for the investigation of seismic

resistance of dams, the MLIT have been made a

periodical collection of the data observed at dams

under its jurisdiction. Now, the dam offices should

report the earthquake records to once a year and the

information of observation system in the event of

new installation or updating of seismographs to the

N1LIM, the MLIT. The detailed format is fixed for

the report and the data in fixed format is sent to the

NILIM by floppy disks.

The MLIT is connecting seismographs installed at dams in a network to construct a system that can

monitor earthquakes in real time for responding to demand of earthquake disaster management. The

dam seismograph network is connected by the direct

microwave circuit, and enables quick collection of

not only the maximum acceleration values but also

time-history acceleration data. This network system

is now monitoring 50 dams of the MLIT.

This system consists of earthquake information

monitoring parts and a data base system Figure 1

1

shows the outline of the network system and Figure

169
12 shows a sample screen on a display terminal.

4.1 Monitoring

The seismograph network monitors earthquakes by

connecting a server at the N1L1M with relay devices

installed at each dam using microwave circuits which

is dedicated to the MLIT.

4.2 Data base

The database consists of basic information of dam,

information of seismograph, earthquake information

and measure data. The following is the actual

examples of the items in database.

a) Basic information ofdam

Dam height, dam length, latitude, longitude,

epicentral distance, dam type, slope gradient

,

etc.

b) Seismograph information

Seismograph model number, specification of

seismograph, installation position, installation

elevation, etc.

c) Earthquake information

Occurrence time, epicenter location, depth,

magnitude, etc.

d) Measure data

Components, peak acceleration, number of

samples, number of data, etc.

4.3 Function and operation

At the instant that a seismograph at a dam detects an

earthquake, the peak acceleration and the measured

seismic intensity of the earthquake are transmitted

through relay devices to the NILIM database. When

the earthquake information arrives at the NILIM

from a dam relay device, a message that an

earthquake has occurred is displayed on a terminal

screen at the same time as an alarm is sounded. And

this system automatically starts the collecting of

earthquake records from dams nearby the dam that

firstly send a data. The dam seismograph network is

connected to other seismograph networks

6

* for

national facilities related rivers and roads, and is

sharing data such as the peak acceleration and

measured SI value etc. The shared information is utilized for immediate

expectation of disaster occurrences such as the

ground liquefaction the damage of bridge etc. Some

kind of digital data such as the peak accelerations

opens public through the Internet. For maintaining of system functions, the state of the

system is constantly monitored once a day by

confirmations of the soundness of each relay device,

and by checking of the existence of untransmitted

earthquake data.

5. FUTURE PLANS

The researches on evaluation methods of seismic

safety of dams have been conducted using data are accumulated in the PWRI and the NILIM.

According as the system to collect the observed

earthquake data from jurisdictional dams has been

completed and networking of seismographs has been

progressing, the related researches are expected to be

advanced. Future plans for the seismograph network

include cooperating with prefectural governments to

link the network to seismographs at all prefectural

dams in order to create a more precise, more highly

concentrated earthquake motion monitoring

environment.

The networking of seismographs enables quick

collection of the earthquake information. The

network is, therefore, counted on to make a big

contribution to supporting an efficient and

appropriate initial emergency action: evaluating the

need for inspections of managed facilities, predicting

the state of damage etc. immediately after the

earthquake when tittle information is available.

It is, moreover, very important to share earthquake

information each other in organizations responsible

to the disaster prevention.

And the collection and accumulation of recordings of

actual seismic data are essential to researches on the

dynamic analysis methods of dams. Therefore, one

of important subjects of the network is linkage to the

170
database of detailed data of foundation and dam

material for the precise estimation of earthquake

motion distribution and dam behavior.

When the networking of seismographs of all dams

underjurisdiction of the MLIT, the collection data by

the NILIM through the network will be take the

place of the report of earthquake data form dam

offices to the NILIM. For improvement of the

reliance of the network system, the supervising

function of system should be strengthened to be

preventing the functional trouble and deterioration.

REFERENCES:

1) Y. Yamaguchi, T. Iwashita: Earthquake Motions

at an Embankment Dam Site and an Estimation

Method of Incident Seismic Waves Using the

Observations, Proceeding of the 32nd Joint Meeting

of UJNR Panel on Wind and Seismic Effects, pp.

253-262, 2001

2) Y. Yamaguchi, T. Iwashita, J. Matsuura:

Seismometers Installed in the Dams Under the

Jurisdiction of the Ministry of Land, Infrastructure

and Transport, Technical Memorandum of PWRI,

No.3768,2001.

3) Y Yamaguchi, T. Iwashita: Earthquake Motion at Rock Foundation and Response of Embankment

Dams by Acceleration Records at Dams in Japan,

Technical Memorandum of PWRI, No.3780, 2001.

4) Fukushima et al.: Shimizu Tech. Res. Bull., Vol.10,

1-11, 1991 (in Japanese)

5) An-naka et al.: 19th JSCE Earthquake Engineering

Symposium, 129-132, 1987 (in Japanese)

6) H. Sugita, T. Hamada: Development of Real-time

Earthquake Damage Estimation System for Road

Facilities, 7th U.S-Japan Workshop on Earthquake

Disaster Prevention for Lifeline Systems, 1997

171
Table 1 Seismograph installation locations

Location Number of Seismographs Ratio (%)

Dam base 400 34.7%

Dam crest 426 37.0%

Free field 290 25.2%

Other 36 3.1%

Total 1152 100.0%

.4 Figure 1 Distribution of dams where seismograph installation is completed

172
Figure 2 Changing number of seismographs

1600

CL| 1200

I

.Sau

u

.oa

800

400

59 The number of analog type seismographs

The number of the digital type seismographs —Total number of the analog type seismographs

- Total number of the digital type seismographs

1200

J3s

cm 900 |

600

300

1

3<*-oa

B

1

On on on t On

NO t- t- 90 00 on 0\ On On On on On On On 7 7 7 7 C 7 7

in o o Wl o NO r- r» OO 00 On On On On On On On On on

o

o

I©o

o

Years

Figure 3 Transition of number of analog type and digital type seismographs

173
Figure 4 Distribution of earthquake epicenters

u<

£

1 riiiiiiiliiiliiiliiijiiiSililiflliiiiiin

0 20 40 60 80 100 120 140 160 180 200

s

-f—t- -4—4.--* i—4-

-H-

. i .... i .... i .... i .... i .... i

.

20 40 60 80 100 120 140 160 180 200

1000

E

100

& 10 Q

mmnnnnnmmmnnnnnmm

r-i.-r v-*t-;-t—,-T"V"^r--*i—•- --•-r---r---T V"i^»»» ^••V% 1 \

1 ^l^iliHirilnillfinl^lnilllillliilli

lEiEEEEEEEEEEEEEJEEEJEEEEEEEEIEEEEiEEEElEEE:

•.-i--t::-t::-t:-.-t:::1:::1:-:.1-.1::::

20 40 60 80 100 120 140 160 180 200

Epicentral Distance (km)

Figure 5 Relationship ofmagnitude, hypocenter depth, and peak acceleration with epicentral distance

174
M5.0~5.9

A M6.0~6.9 • M7.0~7.9

Fukushima(1991)

- -Annaka(1987)

Epicentral Distance (km)

Figure 6 Relationship of epicentral distance and peak acceleration

13 10000

00

O 0. 1

1 1 1 1 1 < 0.01 0.1 1

Period(sec)

Figure 7 Acceleration response spectra (dam base, horizontal)

175
10

isu

u

o-

<u

22 C

§•§

-I<u .S

gi

€ f

1

o

<

o.oi -

0.001

L

— _ 1

' h=0.05

0.01 0.

1

10

Period (sec)

Figure 8 Acceleration response magnification spectra (dam base, horizontal)

176
Period (sec) Period (sec)

1000

100

: h=0.05

.^^.•.•iVl'

i 60-80km

**s» \

V

i \\ *^

i . . A. . ..

IK

117

0.01 0.01 0.1 10

Period (sec) Period (sec)

Figure 9 Acceleration response spectra by epicentral distance interval of 20km

(average, dam base, horizontal)

177
Figure 10 Acceleration response magnification spectra by epicentral distance interval of 20km

(average, dam base, horizontal)

178
I II III

1 1 inn

mini 11 i mil n inn

of LaraL fcaVartroctwo Transport d 10

DVD

Earthquake dairjagc estimation mrk

(previously ill I ill ll)

Figure 1 1 System configuration of the real-time dam seismograph network system

"3 iSSj

jam****

m~*M*

aae-HMf

Mt>«BS miner

IMMi ,

' *ncr.

r SHU

"'«*f

«•»" 3WMK if cat

tit

MHttftft m mr m mTflmrni •m n iilflfcint

a*

li

" U" Mil: »

*»w

*»

*».'. HSHnli

dWHwwaswHHr'" —— yna irt'Mls^irf'W

'

'

*«*»ass»*isa> wmmmmm ~

*ir — it M

—

u m

s -

MM _

«»)»:»».

- ~ - m*&&&*MM ~ izsmsmm win- Hnavnir?

mriwimmm"" " ~ ' mtmmmmm ~'

u

' - ~ uM

am

: M«

MM'

jjjjjjljjlj^^^ mmmmamr~

m^lnimiiai) ' m^mmmi

tmstmftmtm

-a* -

MM;

<&*-t|l*i»mH»«i ,

Figure 12 Example of screen shot of a seismograph network information terminal

179

Cyclically-Induced Pore Pressure at High Confining Stress

by

Michael K. Sharp1 and R. Scott Steedman2

ABSTRACT

Experiments were conducted by the ERDC

Centrifuge Research Team to investigate

effective confining stress effects on liquefaction

potential of fine, clean, Nevada sand, under the

boundary and loading conditions of a centrifuge

model. For this test series, twenty-six level

ground models with either a dense layer over a loose layer or homogeneous profile were tested

in an equivalent-shear-beam box. Some models

were subjected to sequential earthquakes, and

some models were overconsolidated, to observe

stress-history effects on pore pressure

development. The purpose of this paper is to make the initial liquefaction and confining stress studies, centrifuge test procedures, results, and

validation efforts conducted to date known to the

professional community.

KEYWORDS: Centrifuge; confining stress; dams; earthquakes; liquefaction; pore-pressure

1.0 INTRODUCTION

The concern over earthquake resistance of older

embankment dams drives the need for remedial

construction costing hundreds of millions of

dollars of public funds annually. This concern is not generated by actual experience from failures

of well-designed dams but from predictions

based on scientifically sound principles of soil

mechanics. Yet the mechanical response of an embankment dam depends on so many factors

that interact non-linearly that implementation of

these principles in a predictive tool is still as much an art as it is a science. Thus the decisionmaker is faced with the dilemma of making

large investments in the absence of experience

versus protecting the public from the

catastrophic consequences that loss of a major

dam would entail.

Liquefaction is the most critical factor to the

earthquake resistance of embankment dams.

Foundation materials that are structurally ideal

under static loading can be severely weakened

by pore pressures induced from earthquake

motions. Many dams built before awareness of

liquefaction must now be reexamined in light of

potential damage from this threat. An open

question in these reexaminations is the effect of

depth on liquefaction potential. Historically,

there have been no observations of liquefaction

at depths greater than about 30 m or 100 ft (as

summarized in Youd and Idriss, 1997),

suggesting that the high confining stress beneath

most embankment dams wound not be

conducive to significant pore pressure response.

If so, remedial construction could be avoided or

simplified in many dams now under

investigation. Studies to investigate these and

other liquefaction questions relevant to earth

dams are in progress at the U.S. Army Corps of

Engineers, Engineering Research and

Development Center (ERDC), Vicksburg, MS.

Recent findings, focused on level ground stress conditions, are reported in this paper.

2.0 ERDC CENTRIFUGE TESTS

This section describes the test plan, description of

the soils tests, equipment used to conduct the tests, as well as model construction, quality control and

instrumentation. The results and an interpretation

of results conclude the section.

1 Centrifuge Research Center, Engineer Research and Development Center, Vicksburg MS (USA)

2 Whitby Byrd and Associates, London England (UK)

181
2.1 Test Plan

The centrifuge experiments, summarized in

Table 1, were designed to investigate

liquefaction potential of a loose, saturated sand

layer with initial vertical effective confining

stresses ranging from 1 to 10 atm. Level ground

stress conditions were used in all tests. As

shown in Table 1, the models are grouped in

series that correspond to different target ranges

of vertical effective overburden stress near the

bottom of the deposit. All models were shaken

initially at a 50g spin rate with subsequent

shakings at either 50g, 80g, lOOg, or 125g. The

results reported in this paper concentrate on the

first shakings at 50g, unless specifically stated

otherwise. Some models were overconsolidated

by a factor of 2.5 prior to shaking (achieved by

running the centrifuge up to 125g, then back

down to 50g for the shaking event).

Table 2 provides further details on each

experiment, including the densities achieved in

the specimen, the number of earthquake events

and the initial vertical effective stress near the

bottom of the deposit. For the two-layer models,

the target relative density for the upper medium

dense layer was 75 percent while the target

relative density for the lower layer was 50

percent. For the uniform deposits, the target

relative density was 50 percent.

The cross-sections for the different series of

experiment are shown in Figure 1. A variety of

techniques were used to achieve higher

confining stresses while maintaining the same g

level. In some specimens, the water table or

phreatic surface was lowered; in others a surcharge (lead weights) was added on the

surface.

The thickness of the loose sand layer, for the

two-layer case, was constant in all the

experiments (160mm or 8m equivalent field

thickness when tested at 50g). The thickness of

the overlying deposits, and selection of water

table or surcharge were adjusted to achieve the

target effective vertical stress at its mid-depth.

The overall depth of sand within each specimen

was either 300mm or 525mm (15m or 26.25m

respectively, when tested at 50g).

2.2 Soils Used in Centrifuge Models

The Nevada sand used in the models was

characterized by standard laboratory tests to

determine parameters such as dry density and

gradation. The constitutive behavior for Nevada

sand based on laboratory triaxial and simple

tests is given by (Arulmoli et al., 1992). These

tests show the typical reduction with Ka at elevated confining pressure. As a reference, the

Ka relationship for this sand is shown in Figure

2, which clearly displays the effect of confining

pressure. A linear extrapolation of cyclic

strength from low confining pressures would

result in over estimates of the available strength

at high confining stresses.

2.3 Pore Fluid

All models were saturated with a glycerin-water

solution which comprised an 80 percent by

weight mixture of glycerin and water,

determined after study of the sensitivity of the

viscosity to changes in temperature and the mix

proportions, Steedman (1999). The viscosity of

the fluid at lg was selected to maintain proper

scaling of the permeability at 50g. Tables 3 and

4 present key material parameters for Nevada

sand, and the glycerin-water solution

respectively.

The density of the glycerine-water mix was

calculated from as follows.

Pm = pg(mg + mw)/( mg + pg mw)

In this equation pra is the density of the mix, pg is the density of glycerine, mg is the mass of

glycerine, and mw is the mass of water.

2.4 Earthquake Simulation

The earthquake actuator used in the ERDC

experiments is a mechanical shaker designed to

provide a single frequency input motion of

variable duration to the base of the specimen

container. The displacement of the platform is

182
constant at +/- 1mm; the frequency and hence

amplitude is readily changed by varying the

speed of the shaft. The use of a single frequency

of input motion facilitates comparisons with

laboratory tests and practical liquefaction

analysis procedures based on equivalent,

uniform cycles. Table 5 gives the pertinent

characteristics of the input motion for each

model reported in this paper.

2.5 Equivalent Shear Beam Model Container

(ESB)

The soil specimen is built within a hollow

rectangular model container (termed an

equivalent shear beam, or ESB container)

comprised of a series of aluminum alloy rings

stacked one above the other, and separated by an

elastic medium. Several of these chambers have

been constructed, and extensive dynamic

analysis and testing have been carried out to

determine their dynamic response characteristics

(Butler 1999).

The ESB concept is to create an equivalent shear

beam with an average stiffness comparable to the stiffness of the soil specimen during the

initial stages of the shaking, so that the dynamic

response of the box does not significantly

influence the behavior of the soil specimen

inside. A chamber with no stiffness simply adds

mass to the soil specimen, again changing its dynamic response. In experiments involving

liquefaction of large volumes of soil inside the

container, the stiffness (and hence dynamic

response) of the soil changes throughout the

base shaking.

The ERDC ESB model container has internal

dimensions of 627mm deep by 315mm wide by

796mm long. Each of the 11 aluminum alloy

rings was 50 mm high and had a plan area of

0.0675 m

2

. The rings are not stiff enough along

their long dimension to support the outward

pressure from the soil inside under high g, therefore the container relies on vertical reaction

walls, independent of the shaker unit itself, for

lateral support. A rubber sheet separates the

rings from the steel walls on either side, and is

designed to be incompressible but very flexible

in shear. This system has undergone extensive

study to verify that the ESB box is not being

impeded from moving with the shaker platform

even though the container expands during spin

up to 50 g and sits snugly against the reaction

walls.

The end walls of the ESB have thin metal sheets,

termed shear sheets, fixed securely to the base of

the chamber, Steedman et al. (2000). The shear

sheets transmit the complementary shear force

generated by the horizontal shaking on vertical

planes within the specimen to the base of the

container. This improves the uniformity of the

stress field at each elevation along the model,

reducing the tendency for the chamber to rock

thus better representing the stress state boundary

condition that exists in the field.

Butler (1999) conducted a thorough theoretical

and experimental analysis of the dynamic

response of the coupled soil-container system.

At high g the soil and container act as a coupled

system, where the lower stiffness of the

container reduces the natural frequency

(slightly) of the combined system compared to the soil column alone. However, provided the

driving frequency is low relative to the natural

frequency of the coupled system, Butler (1999)

demonstrated that the displacement response of

the system is unaffected compared to the soil

acting independently.

2.6 Model Construction

The specimens were all 300mm wide and

781mm long, between the vertical sidewalls of

the ESB. The specimen was constructed within

the ESB container by pluviating dry Nevada

sand, forming a level sand bed. Instruments

were placed at different depths, during the

pluviation process, to measure pore pressures

and accelerations. The volume and density of

sand poured for each layer was checked to verify

the target relative density. Once complete, the

specimen was fully submerged by introducing

the viscous fluid into the model from the bottom

by gravity over a period of several days.

As the equivalent field depth was increased in

successive experiments, the options were to shake the models at a higher g level (spin rate),

183
to depress the water table or phreatic surface, or

to use surcharging to achieve the target

overburden. It was determined that the models

should all be shaken at 50g, and that surcharging

would be adopted to reach the required stress

levels. In the Model 4 series, two models used a depressed phreatic surface and two used

surcharging. In the Model 5 series, the

equivalent field depth was too great to be

achieved without surcharging. The surcharge

comprised lead strips laid lengthwise along the

surface of the specimen, with additional lead

plate on top where necessary. The phreatic

surface was then maintained at the surface of the

sand bed.

The centrifuge was operated at 50g (for the

majority of tests reported in this paper unless

otherwise specified), creating a field-equivalent

of a site approximately 40 m long by 15 m wide

by over 26 m deep (for the deepest specimens).

Surcharges, variations in the depth of the

phreatic surface, and increased g level were used

to achieve higher effective overburden stresses,

up to 10 arm, as described earlier.

3.0 EXPERIMENTAL RESULTS

In the initial experiment series, Table 2 (model

code 2a - 2f), the behavior of a soil column 15m

deep, with about 1 arm vertical effective stress in

the lower layer was studied. The soil deposit

had a 7m thick upper layer of denser Nevada

sand (relative density of 73 to 83 percent), and a lower layer of looser sand (relative density of 43

to50 percent). The entire model was fully

submerged. Example results for this model

series are shown in Figure 3, for model 2f. This

model has an overconsolidation ratio of 2.5.

The effect of overconsolidation for this series

was to slow the rate of pore pressure increase

throughout the column. Accelerometer records

are shown for depths of 1.5, 3.8, 6.5, 11.3, 14.5,

and 15m. Pore pressure transducer records are shown for depths of 3.5, 6.9, 11, and 14.8m.

The input motions are applied at 0.54 Hz, and

the fundamental period of the soil column is estimated to be 4xl5m/300mps = 0.2 seconds.

Pore pressures begin to rise simultaneously at all 4 transducer depths at 1.5 cycles or 3 seconds of

shaking. During this time, the accelerations are ramping up to 0.1 Og uniformly in the 4 upper

accelerometers, and to 0.1 2g in the input motion

at 15m and the deepest accelerometer at 14.5m.

At 7.8 seconds, acceleration peaks at 0.1 8g for

1.5m, and at 0.15g for 3.8 and 6.5m. From 7-10

seconds, amplitudes decrease slightly to about

0.1 Og in the lower 2 accelerometers and input

motion. At 7 seconds, the rate of pore pressure

buildup increases in the 14.8m transducer, slows

in the 11m transducer, and remains steady in the

6.9m transducer. At 10 seconds, strong shaking

stops in the 1.5m accelerometer, and remains

steady at all other depths. Pore pressures are 100 percent at 1.5m, 35 at 6.9m, 40 at 11m, and

70 percent at 14.8m.

At 26 seconds, pore pressures reach their

maximum values of 100 percent for 6.9m and 95

percent for 11m. Input motions increase from

0.10g to 0.25g, and remain constant at 0.10g at 14.5m. Accelerations decrease to about 0.05g at

3.8, 6.5, and 11.3m. Accelerations reduce to near zero at 1.5m. Accelerations and pore

pressures remain relatively constant at all depths

through the end of shaking at 65 seconds. Little or no pore pressure dissipation is observed

during the 21 seconds since shaking stopped.

More generally, in the model code 2 series, excess pore pressures rose fairly rapidly to 100

percent or near 100 percent of the initial vertical

effective stress (Ru =1) throughout the column

within a few cycles. The acceleration time

histories decrease in amplitude as the Ru approaches 1 . Despite the higher density of the

upper layer, excess pore pressures within it reach high levels.

As the effective vertical confining stress is

increased, the pattern of behavior is initially

repeated. Models 3a-3d in Table 2 were

designed to have an initial vertical effective

stress of 2 arm in the lower, looser layer. All

series 3 models had 18.25m of dense sand,

relative densities of 73-80 percent, on top of 8m

of loose sand, relative densities of 34-54 percent.

All series 3 models were fully submerged.

Example results for this model series are shown

in Figure 4, for model 3c. This model was

normally consolidated. Accelerometer records

184
are shown for depths of 1, 5.5, 14.5, 18.3, 22.3,

25.8m, and the input motion depth of 27m. Pore

pressure transducer records are shown for depths

of 0.75, 5.3, 10.8, 14.4, 18, 22, and 25m. The

input motions are applied at 0.54 Hz, and the

fundamental period of the soil column is estimated to be 4x26.25m/300mps = 0.35

seconds.

The input motions ramped up to a maximum of

0.09g at 3 seconds and dropped down to a steady

0.06g for the rest of cycling. At 25.8m,

accelerations peaked in the 2

nd cycle at 0.14g,

and remained constant at 0.1 2g for the

remainder of cycling. At 22.3m, accelerations

ramped up to 0.09g. Accelerations peaked at 0.035g in 4 seconds at 18.3m. At 5 seconds,

accelerations peaked at 0.035g for 14.5m,

0.022g for 5.5m, and 0.036g for lm. Pore

pressures begin to rise at all depths as soon as shaking starts. There is no lag as observed at 1 arm. At 5 seconds, pore pressures are 100

percent at 0.75m, 80 percent at 5.3m, 70 percent

at 10.8m, 55 percent at 14.4m, and 87 percent at

18, 22, and 25m. Pore pressures reached

maxima of 100 percent in 5 seconds at 0.75m,

16 seconds at 5.3m, 22 seconds at 10.8m, 51

seconds at 14.4m, 7 seconds at 18 and 22m, and

30 seconds at 25m. After shaking stopped, excess pore pressures remained constant at 100

percent.

Above an initial vertical effective stress of

around 250 -300 KPa (3 arm), however, a

different behavior was observed. Under high

effective confining stresses, it was noted that in

many cases, excess pore pressures did not reach

the initial vertical effective stress. Figure 5

shows the results from Model 4d, a deep soil

column with a target vertical effective stress of

500 KPa (5 arm) near the bottom of the model.

Model 4d was subjected to a relatively low level

of shaking amplitude. Nevertheless, in the initial

stages of shaking, the excess pore pressure

development within the loose layer is positive.

Within around 20 cycles, the excess pore

pressure reached 45 percent of the effective

overburden stress. Shortly after this, the rate of

generation of pore pressures all but ceases. There is a slight reduction in the amplitude of

the acceleration, reflecting the increase in excess

pore pressure but the displacement (strain)

cycles continue for a considerable period

without further increase in excess pore pressure.

One concern was that the low amplitude of

shaking might be the reason for this plateau in

pore pressure at only 45 percent of the initial

vertical effective stress. Subsequent models

similar to model 4d were tested with higher

levels of input shaking, revealing the same

pattern as shown in Figure 5.

Figure 6 shows the results of a model very

similar to model 4d, shown in Figure 5, but now

the model is homogeneous rather than a two

layer case. Model 4j had slightly less weights

used for surcharging resulting in an initial

vertical effective stress near the bottom of the

model of around 400 KPa (4 arm). Notice that

the level of input shaking is more than double

that of model 4d, from 0.1 1 g to 0.25 g. Pore

pressure transducers located in the deep loose

layer (33 m and 36 m depths) reach a limit of 78

percent for this case. The largest excess pore

pressure (90 percent) occurs at a depth of 28.5

m.

A second homogeneous model was tested,

without the use of weights for surcharging as was used in model 4j. The results of this test,

model 4k, are shown in Figure 7. Model 4k

depicts a 26 m deep homogeneous deposit of Dr = 49 percent clean sand. The input peak

horizontal acceleration in the model was 0.3 lg.

The model did not contain any lead weight

surcharging; the increase in confining stress was

accomplished by testing the model at increasing

g levels. The results shown in Figure 7 are those

from a 50g test, which produces an initial

effective confining stress near the bottom of the

model of 230 KPa (2.3 arm). As expected, with

a confining stress of only 2.3 arm, the entire

model liquefied. This model was also tested at higher g levels of 80g, lOOg, and 125g, the

results of the testing at lOOg are discussed next.

Figure 8 shows the results of model 4k tested at a centrifugal acceleration of lOOg, which

produces an initial effective confining stress near the bottom of the model of 400 KPa (4 tsf).

The results shown are after the model was

initially tested (input shaking applied) at 50g,

185
held at 50g for several minutes to allow the pore

pressures to dissipate, spun up to 80g and tested,

held at 80g for several minutes to allow pore

pressure dissipation, and then spun up to 100 g

and tested. Pore pressures initially increase

rapidly, and then more slowly, and reach 100

percent or near 100 percent of the vertical

effective stress. At depths of 44.5 and 52 m, the

pore pressures plateau at approximately 82

percent.

The last sets of data are presented in Figures 9

and 10, representing models 5a and 5d

respectively. These results represent two-layer

models surcharged with a large amount of lead

weights and tested at 50g. For model 5a shown

in Figure 9, the confining stress near the bottom

of the deposit is 740 KPa (7.4 tsf). Peak input

acceleration at the base is about O.lg. In this

test, the pore pressure transducers at depths of

50 and 53 m, corresponding to the relatively

loose soil (relative density about 51 percent),

reach a peak value of only about 50 percent, even with numerous additional cycles of

loading. Very little to no excess pore pressure

was recorded in the soil column above 50 m.

Nearly identical results are seen in Figure 10,

model 5d. This model had slightly larger lead

weights resulting in a confining stress near the

bottom of the deposit of 920 KPa (9.2 tsf). The

level of shaking was also increased from that

applied in model 5a from 0.1 g to 0.3 g. At a depth of 63 m, the pore pressure peaked at a value of 55 percent, and at a depth of 59 m the

pore pressure peaked at a value of 40 percent.

Very little to no excess pore pressure was

recorded in the soil column above 59 m.

4.0 DISCUSSION OF RESULTS

In the homogeneous models, excess pore

pressures rise more quickly and to higher levels

throughout the column than in the two-layer

models. Post-shaking pore pressure

redistribution is difficult to see because the

whole column liquefied, so the gradients are

smaller. These observations suggest that pore

pressures are being generated cyclically within

each layer (local generation) in addition to

redistribution effects. The pore pressures do not

reduce within 20 sec after shaking stops, so, dissipation is not observed. Also notice the

sharp correspondence between accelerations and

pore pressure responses. The sharp pore

pressure peaks and liquefaction observations

indicate that sufficient saturation was achieved

with the employed model saturation and

consolidation procedures.

The rate at which PWP is increasing in the

shorter columns appears to be too fast to be

explained by pore pressure redistribution (flow)

alone. Rather, a pressure wave effect may be

occurring in the column. Communication of a high increment of pore pressure originating in a lower layer may be nearly instantaneous

throughout the soil column. Whereas

consolidation requires drainage to take place, the

initial response is essentially undrained, and the

speed of propagation of a compressive wave in

the pore fluid will be many times the velocity of

the shear wave through the soil skeleton. Thus it

is feasible for a pore pressure increment, which

arises during part of a cycle of loading to be

communicated upwards through the soil column

in advance of the shear wave that created it. As

the increments of excess pore pressure tend to be

large at depth, then the potential impact on the

response of the upper layer is increased.

The surcharge has a distinct effect on the

behavior of the soil column. The effect appears

to be even more pronounced in the two layer

models. For the deep two-layer models, excess pore pressures are generated only in the deeper

layers. Dissipation or redistribution effects can be seen in the pore pressure response during and

after shaking stops.

5.0 SUMMARY AND CONCLUSIONS

The relationship between effective confining

stress and liquefaction potential, in terms of

cyclically induced excess pore water pressure,

has been extensively studied in the laboratory

using traditional triaxial, torsional and simple

shear devices. The laboratory studies show that

there is a less-than-linear increase in cyclic shear

strength with increasing effective confining

stress. Laboratory tests indicate that 100 percent

pore pressure response (liquefaction) can occur

186
at high confining stresses in the range 3 to 12

arm. At lower confining stresses, less than 3

arm, stress history strongly influences

liquefaction resistance, K<j relationships, and

threshold cyclic shear strain required to develop

high pore pressures.

Experiments were conducted by the ERDC

Centrifuge Research Team to investigate

effective confining stress effects on liquefaction

potential of fine, clean, Nevada sand, under the

boundary and loading conditions of a centrifuge

model. For this test series, twenty-six level

ground models with either a dense layer over a

loose layer or homogeneous profile were tested

in an equivalent-shear-beam box. Some models

were subjected to sequential earthquakes, and

some models were overconsolidated, to observe

stress-history effects on pore pressure

development. A representative number of the

models tested have been reported in this paper.

6.0 REFERENCES

Arulmoli, K., Muraleetharan, K.K., Hossain,

M.M., and Fruth, L.S., (1992). "VELACS

Verification of Liquefaction Analyses by

Centrifuge Studies Laboratory Testing Program

Soil Data Report." National Science Foundation

Report.

Butler, G.D., (1999). A dynamic analysis of the

stored energy angular momentum actuator used

with the equivalent shear beam container, PhD

thesis, Cambridge University.

Steedman, R.S., (1999). Earthquake engineering

support, Phase 3, Final Technical Report,

N68171-99-C-9021, European Office of the U.S.

Army, London, November.

Steedman, R.S., Ledbetter, R.H., and Hynes,

M.E., (2000). "Liquefaction of sands under high

initial effective confining stresses." Proc.

GeoDenver 2000, ASCE Geotech. Spec. Pub.,

August.

Youd, L. and Idriss, I. M., Editors (1997).

"Workshop on Evaluation of Liquefaction

Resistance of Soils." Proceedings, Salt Lake

City, Technical Report NCEER-97-0022,

sponsored by FHWA, NSF and WES, published

by NCEER

TABLE 1. Outline summary of centrifuge model test program

Model

series

Models in

series

Effective

overburden

stress near bottom

Equivalent

field depth

(approx)

Depth of

specimen

Notes (all specimens constructed

from Nevada Sand, tested at 50g

unless indicated)

2 a,b, c, d, e, f 1 tsf 15 m 300 mm Fully submerged

3 a, b, c, d, e 2 tsf 26 m 525 mm Fully submerged

4

a, b, c, d 3 - 5 tsf 26 -40 m 525 mm

Lowered water table or surcharge

(weights)

e, f, g, h, i, j 3.9 -9.7 tsf 30 - 60 m 525 mm

Surcharge (weights) and change in g

level, fully submerged

k 1.9 -4.8 tsf 25 -39 m 525 mm

Change in g level, no surcharge, fully

submerged

5 a, b, c, d, e 7-10 tsf 54 - 63 m 525 mm Surcharge (weights), fully submerged

187
TABLE 2. Summary description of centrifuge model test program

Model

Code

Dpth.

(mm)

Relative Density

(Dr)*

hottftm nf* deoosit (tsf)

OCR Number

of Eqks.

V^UlllIllCll la • All UlUUClj lUllall UllCU

from 1 1 Will lVpvsiHa ll^TdUA SanH kJullU and MllU fpcfprl ICSltU (11 at SOo* JUc

unless specified**

Za JUL) A A 0/ 1 A. A 1-1 A O ^ 0/ Jn„n A 44% loose, od /o dense

-j -j

5 Fully submerged

ZD inn jUU ju/o loose, /j /o dense — — z Fully submerged

zc inn 4y /o loose, /*f /o dense — —

cJ ruuy suDmergea

Za inn C AO/ Irt^oia *7^0/ si a jUyo loose, OYo dense —

A Fully submerged

ze inn jUU /i AO/ 1«ACO '710/ i-lonco 4y /o loose, / j /o aense — Z.J AH ruuy suumergea

ZI inn jUU CAO/ 1 AACO *7 CO/ i-lonoo ju/o loose, /j/o dense Z.J A ruuy suomergea

ja JZJ

1/10/ 1 AAPD *710/ /lanco -?4 /o loose, / j /o dense Z Z

Lj ii 1 ITT oil E"\ r^"k £2* v t~x A/~

i

ruuy suomergea

3b 525 49% loose, 77% dense 2 1 3 Fully submerged

3c 525 /< Af)/ 1 *700/ J — —. — » 49% loose, 79% dense 2 3 Fully submerged

3d 525 54% loose, 80% dense 2 2.5 4 Fully submerged

4a 525 a t\n / 1 n An/ J 49% loose, 80% dense 3 4 Saturated to top of loose layer only.

4b 525 56% loose, 74% dense 3 2.5 4 Saturated to top of loose layer only. 1

4c 525 50% loose, 75% dense 4.7 4 Fully submerged, surcharge (weights)

4d 525 50% loose, 68% dense 4.7 2.5 4 Fully submerged, surcharge (weights)

4e 525 47% uniform 3.9-7.8

2,1,1** Fully submerged, surcharge (weights),

Shaking at 50, 80, lOOg

4f 525 55% uniform 3.9-9.7 1,1,

2,1**

Fully submerged, surcharge (weights),

**Shaking at 50, 80, 100, 125g

4g 525 50% uniform 3.9-9.7 1

1,1,

1,2** Fully submerged, surcharge (weights),

**Shaking at 50, 80, 100, 125g ;

4h 525 50% uniform 3.9-9.7 I 1,1,

1,1**

Fully submerged, surcharge (weights),

**Shaking at 50, 80, 100, 125g

4i 525 50% uniform 3.9-9.7 1 1,1,

1,1**

Fully submerged, surcharge (weights),

akakOI. 1 * _ j_ c f\ OA 1 AA 1 ^ C **Shaking at 50, 80, 100, I25g

4j 525 50% uniform 3.9-9.7 1 l,l,

f~* 11 V J i_ / • 1_ , \ Fully submerged, surcharge (weights),

**Shakine at 50 80 100 I25e

4k 525 50% uniform 1.9-4.8 1,1,

1,2**

Fully submerged, no surcharge,

**Shaking at 50, 80, 100, I25g.

5a 525 51% loose, 72% dense 7.4 4 Fully submerged, surcharge (weights)

5b 525 49% loose, 76% dense 7.4 2.5 4 Fully submerged, surcharge (weights)

5c 525 52% loose, 75% dense 9.2 3 Fully submerged, surcharge (weights)

5d 525 57% loose, 80% dense 9.2 1 Fully submerged, surcharge (weights)

5e 525 50% uniform 8.4 7 Fully submerged, surcharge (weights)

*Relative density of the portion of the model termed 'loose', always refers to a 160 mm thick layer at the bottom of

the model. Relative density of the portion of the model termed 'dense', refers to that portion of the model above the

'loose' layer (ranges in thickness from 140 to 365 mm)

188
Table 3. Nevada Sand Properties (as

measured)

Specific Gravity 2.64

Maximum void ratio

0.757 (density

1502.53 kg/m3 ) Minimum void ratio

0.516 (density

1741.21 kg/m3 ) D50 0.18 mm

Dio 0.11 mm

Table 4. Pore Fluid Properties (as

measured)

Density 1200 kg/cm3

Viscosity 50 cs

Specific Gravity 1.26

Composition

85% glycerine, 15%

water (by weight)

Table 5. Characteristics of Input

Motion

Model

Freq.,

Hz

Number

of cycles

2f (Figure 3) 0.54 0.25 34

3c (Figure 4) 0.54 0.09 35

4d (Figure 5) 0.54 0.11 50

4j (Figure 6) 1.18 0.25 57

4k(50g),

(Figure 7)

1.17 0.32 57

4k(100g),

(Figure 8)

0.88 0.42 85

5a (Figure 9) 0.54 0.10 35

5d (Figure

10)

0.86 0.32 51

189
796

Lead surcharge (except Model 4k) __ ^y.,..V.;,. y,|

-796-

Nevada sand RD = 75%

T Fully

submerged

Nevada sand, RD = 50% ; . v. 160

525

Typical cross section through Model series 2

796

Typical cross section: Models 4 e, g, h, i, j, k and 5e

796 1

Fully submerged

Nevada sand, (Model 3)

Rj3 = 750/,, 300 mm depth

(Models 4a, b) 2

I

-_£CCC<XXXS'i >':' £ Nevada sand, RD= 50%

365

ljO

Lead surcharge

Nevada sand,

RD = 75%

Fully

submerged

Nevada sand, RD = 50%

365

160

Typical cross section: Models 3 and 4 a, b Typical cross section: Models 4 c, d, f and 5 a-d

Figure 1 . Cross sections through different centrifuge models (all dimensions in mm)

1.4

1.2

1

0.8

0.6

0.4

0.2

0

0 50 100 150 200

Vertical Effective Confining Stress, kPa

Figure 2. Ka relationship for Nevada sand based on cyclic triaxial test data as reported by

Arulmoli etal., 1992.

190
&0i

i

o

o

_IVES_j

1 1 1

1

Soil Depth = 3.5 m

"

i i

i i i i ^A*VWWWWVVVVVVVVy^ ;

Soil Depth = 6.9 m

iii- i

" IVES

1 1 1

Soil Depth = 1 1 m

20 40 60 80 100

50g

i

Time, sec

-796mm

40 60

Time, sec 'i^j Nevada Sand, i^i;

jljjjj Dr = 75% jjjjlj

Nevada Sand,

D =50%

T

140mm

160mm

& Acc

• PPT

Figure 3. Results from model 2f showing recorded accelerations and excess pore pressure

191
o

o<

0.12 0.08 0.04

0.00

-0.04

-0.08

-0.12

365mm

160mm

•ure 4. Results from model 3c showing recorded accelerations and excess pore pressure

192
0.15 0.10

0.05 0.00

-0.05 -0.10

-0.15

400

300

200

100

0

-100

500

400

300

200

100

0

-100

IVES

- IVES

Oh

22O

Oh

(/}

</} ua

. IVES

IVES

500

400

300

200

100

0

-100

500

400

300

200

100

0

-100

600

500

400 h IVES

300

200

100

0

-100

600

500

400

300

200

100

0

-100

Soil Depth = 5.2 m

Equivalent Depth = 19 m

Soil Depth = 11.2 m . Equivalent Depth = 25 m

Soil Depth = 14.2 m

Equivalent Depth = 28 m

Soil Depth = 18.2 m

Equivalent Depth = 32 m.

Soil Depth = 22.2 m.

Equivalent Depth = 36 m1

i i i

- IVES Soil Depth = 26.2 m

i Equivalent Depth = 40 m.i i i 0 20 40 60 80 100 120

Time, sec

50g

_ 796mm i

20 40 60 80 100 120

Time, sec

1 LeaJ Weighty i

Nevada Sand,

D=68%

Nevada Sand,

D =49%

365mm

160mm

cc PPT Figure 5. Results from model 4d showing recorded accelerations and excess pore pressure

193
525mm

Figure 6. Results from model 4j showing accelerations and excess pore pressures.

194
a"o

1

V

o

o<

40 60

Time, sec

80

1

Soil Depth = 2 m -

i i

1 1

Soil Depth = 11.3 m I

50g

„. 40 60 80

Time, sec

-796mm

lit

Nevada Sand,

;

Dr = 49%

525mm

• Acc

• PPT

Figure 7. Results from model 4k, 50 g showing accelerations and excess pore pressure.

195
1 1

Soil Depth = 3.5 m

120

100

80

60

40

20

0

-20 200

150

100

50

0

-50

IVES

I—r Soil Depth = 11.3 m

1 1 "IVESi i i i i

i i SoiJ Depth = 22r 5 m

"

500

400

300

200

100

0

-100

1

1

1 1 1

; 1

1 1 Soil Depth = 52 m -

i i

20 40 60 80 100 120 140 160

Time, sec

20 40 60 80 100 120 140 160

Time, sec

796mm

ipog

f

; Nevada Sand,

! D=49%

525mm

• Acc

• PPT

Figure 8. Results from model 4k, lOOg showing accelerations and excess pore pressures.

196
60d

oa

u

oo<

0.10

0.05

0.00

-0.05

-0.10 0.10

0.05

0.00

-0.05

-0.10

0.10

0.05

0.00

-0.05

-0.10

0.15 0.10

0.05 0.00

-0.05 -0.10

-0.15

Equivalent Depth = 29 m

Equivalent Depth = 39 m

Equivalent Depth = 42.5 m

11.2 m

th= 18m - fc

T 1

Soil Depth = 22.3 ifi

Equivalent Depth = 50 m

20 40 60

Time, sec

100

a

S"

lO

C/5 OJ

o

COu

oXW

Soil Depth = 5.5 m

Equivalent Depth = 33 m

Soil Depth = 11.2m

Equivalent Depth = 39 m

Soil Depth = 14.8 m

Equivalent Depth = 42.5 m

Soil Depth = 18 m

Equivalent Depth = 46 m

Soil Depth = 22.3 m

Equivalent Depth = 50 mj . \ i

Soil Depth = 25.3 m

Equivalent Depth = 53 m40 60

Time, sec

Y Lead Weight

7

80 100

-796mm

1

Nevada Sand,

D =72%

Nevada Sand,

D =51%

365mm

160mm

Acc PPT

Figure 9. Results from model 5a showing accelerations and excess pore pressure.

197
~l 1 1

Soil Depth = 1 m

uivalent Depth = 38 nr

= 5.5m

Equivalent Depth = 42 m'

1000

800

600

400

200

0

-200

IVES Soil Depth = 5.5 m

Equivalent Depth = 42 m

Soil Depth = 11.2nr

Equivalent Depth = 48 m

1000

800

600

400

200

0

-200

IVES Soil Depth = 11.2 mEquivalent Depth = 48 m'

60

o1

O

O

<

03

40 60

Time, sec

1000

800

600

400

200

0

-200

1000

800

600

400

200

0

-200

1200

1000

800

600

400

200

0

-200

1200

1000

800

600

400

200

0

-200

IVES

IVES

IVES

Soil Depth = 14.5 m

Equivalent Depth = 52 m

Soil Depth = 18.3 m

Equivalent Depth = 55 m

Soil Depth = 22 inEquivalent Depth = 59 m1

- IVES

i i i

Soil Depth = 25.8 m

;

Equivalent Depti) = 63 m20 40 60

,„ Time, sec

50g

Y l-ead Weight

7

80 100

-796mm

Nevada Sand,

Dr =80% |j§!

Nevada Sand,

D =57%

365mm

160mm

• Acc PPT

Figure 10. Results from model 5d showing accelerations and excess pore pressures.

198
SESSION 5

TRANSPORTATION SYSTEMS

An Evaluation of the Force Reduction Factor in the Force-Based Seismic Design

by

Gakuho Watanabe 1 and Kazuhiko Kawashima1

ABSTRACT

This paper presents an analysis of the force

reduction factors used in the force-based seismic

design of structures. The force reduction factors

are evaluated based on 70 free-field ground

motions. Scattering of the force reduction factors

depending on ground motions and the effect of

damping rations assumed in linear and nonlinear

responses are clarified. A new formulation of the

force reduction factors is presented.

KEY WORDS:

Seismic design, Forced-based design, Response

modification factor, Force modification factor,

Seismic response

1. INTRODUCTION

In the force based seismic design, it is usual to

estimate the demand from a linear response of a

structure by dividing it by the force reduction

factor. The force reduction factor or response

modification factor, which is often called q-factor

or R-factor, has an important role in the estimation

of design force of a structure. An early study by

Newmark and Hall (1973) revealed the fact that

the equal displacement assumption and the equal

energy assumption provide a good estimation of

the force reduction factors at long and short

periods, respectively. This affected an important

effect to seismic design criteria worldwide.

Various researches such as Nassar and Krawinkler

(1991) and Miranda and Bertero (1994) have been

conducted on the force reduction factors. In

particular, Miranda and Bertero provided a

detailed review on the force reduction factors.

However, in spite of the importance in seismic

design, less attention has been paid to large

scattering of the force reduction factors depending

on ground motions. Since the scattering is so large,

only the mean values of the force reduction factor

is not sufficient to evaluate a force reduction

factor for design. Assumption of damping ratio for

evaluating the linear and nonlinear responses is another important point. Although it has been

general to assume the same damping ratio for the

linear and nonlinear responses, it depends on how

the force reduction factors are used.

This paper present an analysis on the force

reduction factors based on 70 free-field ground

motions. The scattering of the force reduction

factors depending on ground motions and the

effect of assumption of damping ratios in linear

and nonlinear responses are clarified.

2. DEFINITION OF FORCE REDUCTION

FACTOR

If one idealizes a structure in terms of a single-degree-of-freedom (SDOF) oscillator with an elastic perfect plastic bilinear hysteretic

behavior as shown in Fig. 1, the force reduction

factor RM may be defined as

'Tokyo Institute of Technology, O-Okayama, Meguro, Tokyo, Japan, 152-8552

201
RH

(T, Ht ' %EL ' 4nl ) (1)

_ 1 AW

(3)

in which T : natural period, Fr

L and FyL

: maximum restoring force in an oscillator with a

linear and a bilinear hysteresis, respectively, //r

:

target ductility factor, and %EL and gNL : damping ratio assumed in the evaluation of linear

and bilinear responses, respectively. The natural

period T may be evaluated based on the cracked

stiffness of columns. Representing u

y the yield

displacement where the stiffness changes from the

cracked stiffness to the post-yield stiffness, a

target ductility factor pT may be defined based

on the yielding displacement u

y as

(2) u ,

in which umaxT is a target maximum

displacement of an oscillator. The post-yield

stiffness is assumed to be 0 in the present study.

Since the damping controls structural response,

it has to be clarified carefully. A structure under a

strong excitation generally exhibits strong

hysteretic behavior, and this results in an energy

dissipation in a structure. For example, the

flexural inelastic deformation of columns

contributes to energy dissipation in a bridge.

Hence, the evaluation of damping ratio depends

on the idealization of such an energy dissipation.

If one idealizes the energy dissipation in nonlinear

structural components by incorporating nonlinear

elements that represent the hysteretic behavior, the

energy dissipation in the nonlinear structural

components is automatically included in the

analysis. On the other hand, if one idealizes the

nonlinear structural components by elastic linear

elements, the energy dissipation in the nonlinear

structural components has to be included in the

analysis by other means. The equivalent viscous

damping ratio %h

is generally used for such a purpose as

4/r W

in which AW and W represent an energy

dissipation in a hysteretic excursion and the

elastic energy, respectively. For example, in an

oscillator with an elastic perfect-plastic bilinear

hysteresis, the equivalent damping ratio %h

is 2 fi-\

(4)

Fig. 2 shows the equivalent damping ratio by

Eq. (4). It is generally very large such as 0.4 at the

target ductility factor of 3-5.

In addition to such hysteretic energy

dissipation, there must be some other sources of

energy dissipation (for example, Kawashima,

Unjoh, Tsunomoto 1993). The radiation of energy

from a foundation to surround ground contributes

to energy dissipation. Structural damping such as

friction at connections may be important in many

structures (for example, Kawashima and Unjoh

1989). Viscous damping due to friction with air is

generally predominant in a structure with a long

natural period. It is general to idealize those

sources of energy dissipation in terms of the

equivalent viscous damping.

If one considers a structure in which the

flexural hysteretic energy dissipation is predominant with other sources of energy

dissipation being a secondary importance, the

total damping ratio £ of a SDOF oscillator

may be provided as 4eq=Zh+Zoth . (5)

in which ^his the damping ratio that accounts

the hysteretic energy dissipation by Eq. (3), and

%oth is the damping ratio that accounts the energy

dissipation other than the hysteretic energy

dissipation.

In the evaluation of the force reduction factor

RM

based on Eq. (1), how damping ratios are assumed in the evaluation of the linear and the

nonlinear responses is important. If one assumes

the damping ratios as

202
%EL=Zeq

^

%NL=Zoth (6)

the energy dissipation is essentially the same

between the linear and the nonlinear responses.

Hence, the force reduction factor by Eq. (1)

represents the difference of restoring force

between the linear and nonlinear responses. Thus,

Eq. (1) reflects the effect of nonlinear response of

an oscillator.

On the other hand, if one assumes the damping

ratios as

eq (7)

or,

the force reduction factor by Eq. (1) includes the

effect of different energy dissipation between the

linear and nonlinear responses, in addition to the

effect of nonlinear response. By assuming Eq. (7)

in the evaluation of nonlinear response, the

hysteretic energy dissipation in the nonlinear

structural components is counted by the

equivalent viscous damping in addition to the

inelastic excursion in the nonlinear elements. As a

consequence, the hysteretic energy dissipation in

the nonlinear structural components is counted

twice in the evaluation of nonlinear response. On

the other hand, if one assumes Eq. (8), the

hysteretic energy dissipation is not taken into

account in the evaluation of linear response.

It should be noted here that which is

appropriate among Eqs. (6), (7) and (8) depends

on how the force reduction factor is used. Based

on the original definition inherent to the force

reduction factor, it seems that Eq. (6) is the most

appropriate. Eq. (8) generally provides

conservative estimation for the force reduction

factors. If ^oth tne difference of the force

reduction factors among Eqs. (6), (7) and (8) is

limited. It should be noted that gh , E,oth and £

depend on the type of a structure, mode shape,

hysteresis and the target ductility factor.

Although the equivalent damping ratio ghis

very high as shown in Fig. 2, it is not general to

assume such a high damping ratio in seismic

design of a bridge structure. It is because a bridge

structure is generally more complex than a SDOF

oscillator, and this makes the relative contribution

of the hysteretic energy dissipation of columns

less significant. Since it is general practice in a standard bridge structure to assume about 0.05 for

the damping ratio including hysteretic energy

dissipation of columns, £ is assumed to be 0.05

in the present study based on Eq. (6). Hence, it is assumed here that £h and %oth is 0.03 and 0.02,

respectively.

Using those damping ratios, the force

reduction factors are evaluated in this study based

on Eq. (6). However, an analysis assuming Eqs.

(7) and (8) is also conducted for comparison with

the previous studies.

3 . REVIEW OF THE PAST

INVESTIGATIONS

An early study for the force reduction factor was

conducted by Newmark and Hall (Newmark and

Hall 1973). They used 10 ground motions

recorded in the 1940 Imperial Valley Earthquake.

They assumed %EL = ^NL =0.05, and proposed a

force reduction factor as

l

V2//-1

{§<T<TX /10)

2.5131og(l/V2//-l)

47,1 )

(7j /10<7<7! I A)

V2y"-1 (7j/4<7<7]')

7///7j (7/ <7<7,)

M (7i <7<72 )

(9)

where,

7, = 2n

tea A

203
[ 1 ='l

T =2n <tedR

(10)

in which, A , V and D represent peak ground

acceleration, velocity and displacement,

respectively, and <j>ea , <f>ev and <f>ed represent

the amplification for acceleration, velocity and

displacement, respectively.

Nassar and Krawinkler proposed a force

reduction factor, assuming ^EL = gNL =0.05,

based on an analysis for 15 ground motions as

(Nassar and Krawinkler 1991)

/^={c(//-l) + l}

,/c

(11)

where

c(T,a)

b

H (12)

l + T

a T

in which a represents a ratio of the post-yield

stiffness to the initial elastic stiffness, and a and

b are coefficients depending on a . Nassar and

Krawinkler precisely analyzed the effect of

stiffness deterioration, and provided the

coefficients a and b depending on a

.

Miranda and Bertero proposed a force

reduction factor, assuming %EL = =0.05,

based on an analysis for 124 ground motions as (Miranda and Bertero 1994)

RU =

M ®(T,Tg

) + 1>1 (13)

where,

1 +

1 \QT-fiT IT—1 exp

3 f 3

\nT —

5

\2

(rock)

1 +

1 \2T-fiT 5T

exp

f i \ 2

InT--

5

T 3T

1 + — ^-exp

3T AT

-3

(alluvium)

\2~

(soft)

(14)

in which T represents a most predominant

period.

It has been known that the equal energy

assumption provides a good estimation for the

force reduction factor at short periods while the

equal displacement assumption at long periods.

The force reduction factor provided by the equal

energy and the equal displacement assumptions

are given as

R (equal energy)

/?„ = ju (equal displacement)

(15)

(16)

Application of Eqs. (15) and (16) and a comparison of the present study to the previous

models will be described later.

4. FORCE REDUCTION FACTOR FOR

BILINEAR OSCILLATORS

Force reduction factors were evaluated for target

ductility factor juTof 2, 4, 6 and 8 assuming an

elastic perfect-plastic bilinear hysteresis. Damping

ratio in the linear and nonlinear analyses is assumed as ££L =0.05 and gNL =0.02 based on

Eq. (6). Seventy free field ground accelerations by

64 shallow earthquakes with depth less that 60 km

were used for analysis. They are classified into

three soil conditions depending on the

fundamental natural period of subsurface ground

204
Tg

; stiff (T

g

< 0.2 s), moderate (0.2 <Tg < 0.6 s)

and soft (T

g

>0.6 s) (Japan Road Association

2002). Number of records in the stiff, moderate

and soft categories is 16, 39 and 15, respectively.

Distribution of peak ground accelerations on the

earthquake magnitudes and epicentral distances is shown in Fig. 3. The peak accelerations are in the

range of 0.1-8m/s

2

, and the epicentral distances

are in the range of 10-500 km.

Fig. 4 shows the force reduction factors for the

70 ground motions. Only the results for /iT

=4

and 6 are presented here since the results for other

target ductility factors show the similar

characteristics. It is seen in Fig. 4 that scattering

of the force reduction factors depending on

ground motions is significant. For example at

natural period of 1 second, the force reduction

factors varies from 1.9 to 10.3 depending on

ground motions for //7-=4 at the moderate soil

sites. It is apparent that such a large scattering of

the force reduction factors result in a large change

of sizing of a structure in seismic design.

Obviously smaller force reduction factors should

be assumed in design to provide conservative

design. It is observed in Fig. 4 that the

dependence of force reduction factors on the soil

condition is less significant. This will be discussed

later.

Since the scattering of the force reduction

factors depending on ground motions is so large

that the means +/- one standard deviations of the

force reduction factors were obtained for each

target ductility factor, natural period and soil

condition. Fig. 5 shows the mean values and the

mean values +/- one standard deviations of the

force reduction factors presented in Fig. 4. The

force reduction factors predicted by Eqs. (15) and

(16) based on the equal displacement and the

equal energy assumptions are also presented here

for comparison. The mean values of force

reduction factors increase as the natural periods

increase, and then they approach to nT

at long

period. It has been pointed out in the previous

researches that the Eq. (15) provides a good

estimation to the force reduction factor. However,

it is noted that Eq. (15) provides a good estimation

to the mean values, but it considerably

underestimates the force reduction factors

corresponding to the mean values minus one

standard deviations. On the other hand, Eq. (16)

provides better estimation to the mean values

minus one standard deviations. Taking account of

the force reduction factors having considerable

scattering depending on ground motions, it seems

reasonable to consider a certain redundancy in the

estimation of the force reduction factor in design.

Based on such a consideration, it is more

conservative to assume Eq. (16) instead of Eq.

(15) for a design purpose.

Fig. 6 shows the dependence of the standard

deviations of force reduction factors cr(/?..) on

the natural periods T and the soil condition.

Similar to the mean values, the standard

deviations a(RM

) increase as the natural periods

increase, and decrease after taking peak values at

natural period of 1-2 second. Fig. 7 shows the

dependence of the standard deviations cr(/?

/i ) on

the target ductility factors jUj . The standard

deviations cr(RM

) increase as the target ductility

factors increase. The relation may be

approximated by a least square fit as

-0.328 + 0.379 //r

(stiff)

a(RM

) = < -0.292 + 0.378 fiT

(moderate) (17)

-0.354 + 0.409 nT

(soft)

As the soil condition dependence of <j(Rm

)

is less significant as shown in Fig. 7, Eq. (17) may

be approximated as

cr(/?

// )*-0.3 + 0.4-//7

- (18)

5. FORMULATION OF FORCE REDUCTION

FACTORS

To idealize the mean values of the force reduction

205
factors in Fig. 5, they are represented as

where,

(19)

(20)

in which a , b and c are parameters to be

determined.

Since Ru = fi at T - a in Eq. (20), the

parameter a represents the period where RMis

equal to // (Point P) as shown in Fig. 8. Because

the gradient of RMis "'-^y-ff-** <2.)

dT MT-a)

it is c-(^-l) at Point P. Consequently, the

parameter c represents the gradient at Point c

divided by ju - 1 . Representing Q as the point

where Ru

takes the peak value, Mb represents

the period between Points P and Q.

Based on the definition, the following

condition has to be satisfied in R„

lim Rn - 0

r-»o (22)

Hence, the coefficient c can be eliminated as

c = \l ae

ab

(23)

Substitution of Eq. (23) makes Eq. (20) as

ae

(24)

It is noted that Eq. (19) automatically satisfies

the following condition

lim/?„ = fx =M (25)

7"-»oo

It is a feature of the above formulation that the

equation is simpler and the physical meaning of

the parameters a and b is clearer than the

previous studies.

The mean values of force reduction factors in

Fig. 5 were fitted by Eq. (19) using a nonlinear

least square method (Press et al 1996). Table 1

represents the a and b as well as the

regression coefficients. Although the regression

coefficient is not high enough for some

combinations such as fiT =2 and stiff sites, it may

be accepted in other conditions. As shown later,

the fitting is not necessarily poor for a combination of /jt =2 and stiff sites.

Fig. 9 shows parameters a, lib and a + l/b.

Parameter a is in the range of 1.0-1.4 second at

stiff and moderate sites, and 1 .5-2.4 second at soft

sites. They are less sensitive to the target ductility

factor nT between 2 and 8. As described before,

a represents the period where Ru = ju, which

implies that the equal displacement assumption by

Eq. (16) provides the best estimation at period a . Consequently, the accuracy of equal displacement

assumption is high at 1.0-1.4 second at stiff and

moderate sites, and 1 .5-2.4 second at soft site. As shown in Fig. 8, a + l/b represents the

natural period where R

p

takes the peak value. It

is 1.5-2 second at stiff and moderate sites, and

2.5-3.5 second at soft site. It slightly increases as

target ductility piT

increases.

The natural periods where R^ take the values

predicted by Eq. (15) based on the equal energy

assumption are obtained as shown in Fig. 10.

They are in the range of 0.2-0.36 second, 0.26-0.4

second and 0.4-0.6 second at stiff, moderate and

soft sites, respectively. They are much shorter than

the natural periods where the equal displacement

assumption provides the best approximation.

Fig. 11 compares the mean force reduction

factors presented in Fig. 5 to the values predicted

by Eq. (19). Although some discrepancies are observed at larger target ductility factors, Eq. (19)

provides a good estimation for the mean force

reduction factors.

Fig. 12 shows the effect of soil condition on the

mean force reduction factors estimated by Eq.

(19). The effect of soil condition is less significant

on the force reduction factors, in particular at

small target ductility factors.

As shown in Fig. 4, scattering of the force

206 !

i
reduction factors around the mean values is

extensive. Hence, the force reduction factors

corresponding to the mean values m substituted

by a standard deviation cr(^) are evaluated as shown in Fig. 13. The mean and the standard

deviation of force reduction factors were

evaluated by by Eq. (19) and Eq. (18),

respectively, in this estimation. They are of course

close to the force reduction factors of the mean

minus one standard deviation directly computed

from the 70 ground motions (refer to Fig. 5). The

force reduction factors predicted by Eq. (15)

based on the equal energy assumption are

presented here for comparison. From Fig. 13, it is

seen that at /uT =4, the equal energy assumption

provides a good estimation at natural periods

longer than 0.5 second at stiff and moderate sites

and 1.2 second at soft sites, while it provides

underestimation at natural periods shorter than

those values. On the other hand, at juT

=S, the

equal energy assumption provides a good

estimation at 0.6 second at stiff and moderate sites

and 1 second at soft sites. It underestimates and

overestimates the force reduction factors

corresponding to the mean minus one standard

deviation at natural periods shorter and longer,

respectively, than the above natural periods.

6. EFFECT OF DAMPING RATIOS

In the preceding analysis, the force reduction

factors were evaluated based on Eq. (1) assuming

^££=0.05 and ^NL =0.02. However in the past

researches, damping ratios were usually assumed

as %EL =

%NL =0-05. Consequently, the same

analysis presented in the preceding chapters was

conducted by assuming £el =

<?NL =®-®5 based

on Eq. (7) using the same ground motion data set. For comparison, an analysis was also conducted

assuming %el = <Znl =002 based on Eq. (8).

Tables 2 and 3 show the parameters a and b

determined for a combination of %el =

%nl =0-02

%EL =%NL =0-05, respectively. Fig. 14

compares a and a + l/b thus determined. Also

presented in Fig. 14 are a and a + l/b used in

the preceding chapter (^EL =0.05 and gNL =0.02,

refer to Fig. 9). It is seen in Fig. 14 that both a

and a + l/b at the same target ductility factors

are the shortest for a combination of

%El = =0.02 and the longest for a combination of %EL =0.05 and %NL =0.02.

Parameters a and a + \lb for a combination of

%EL =£m, =0-05 are between the two cases.

Fig. 15 compares the mean values of the force

reduction factors based on the three assumptions

of damping ratios. Original force reduction factors

computed from the 70 ground motions are also

presented here for comparison. A systematic

difference of the force reduction factors is observed reflecting the dependence of a and

a + l/b on the damping rations. The combination

°f 4el =

<?nl =0-02 provides the largest

estimation for the force reduction factors, while

the combination of %EL

=0.05 and 4nl =0-02

provides the smallest estimation. The combination

°f %EL =

4nl =0.05 provides the estimation

between the two cases.

7. COMPARISON WITH THE PREVIOUS

STUDIES

Fig. 1 6 shows a comparison of the force reduction

factor in the present study by Eq. (19) to Nassar

and Krawinkler by Eq. (11) and Miranda and

Bertero by Eq. (13). Since it is assumed in Eqs.

(11) and (13) that %EL = %NL =0.05, the same

damping ratios are assumed in the present study

for comparison. The original mean values of the

force reduction factors computed from the 70

ground motions are also presented here for

comparison. It is noted that definition of soil

conditions is not the same among three researches.

Hence they are classified into stiff, moderate and

207
soft. In the Miranda and Bertero formulation, T

g

was assumed 1.5 second at soft (alluvial) site in

Eq. (14).

From Fig. 16, it is seen that the present study

provides a quite similar result to the formulations

by Miranda & Mertero and Nassar & Krawinkler

if the same damping ratios are assumed in the

evaluation of linear and nonlinear responses.

8. CONCLUSIONS

An analysis was conducted for the force reduction

factor based on response of SDOF oscillator using

70 free-field ground motions. Based on the

analysis presented herein, the following

conclusions may be deduced:

1) A new formulation as shown in Eqs. (19) and

(24) was developed. The formulation is simpler

than the past formulations. Parameters a and

a+l/b express the natural period where is

equal to /u and RM

takes a peak value,

respectively.

2) Difference of the damping ratios assumed in

the evaluation of linear and nonlinear responses

(<^EL md provides a systematic difference

in the force reduction factors. The combination of

^££ = ^^=0.02 provides the largest estimation

for the force reduction factors, while the

combination of %EL =0.05 and i; NL =0.02

provides the smallest estimation. The combination

of = gfljL =0.05 provides the estimation

between the two cases. Hence, the damping ratios

have to be carefully assumed keeping how the

force reduction factors are used in mind.

3) Scattering of the force reduction factors

depending on ground motions is significant.

Although it has been pointed out that the equal

displacement assumption by Eq. (15) provides a good estimation to the force reduction factors, it

provides a good estimation only to the mean

values; however, it considerably underestimates

the mean minus one standard deviation. On the

other hand, the equal energy assumption by Eq.

(16) provides a better estimation to the force

reduction factors corresponding to the mean

minus one standard deviation, although it provides

too conservative estimation to the mean values.

Taking account of the considerable scattering of

the force reduction factors depending on ground

motions, it is conservative to assume the equal

energy assumption instead of the equal

displacement assumption.

4) The response modification factors in the

present study by Eqs. (19) and (24) provides quite

close force reduction factors proposed by Nassar

and Krawinkler, and Miranda and Betero, if the

damping ratios are assumed as %EL = ^NL =0.05.

REFERENCES:

Japan Road association (2002), Part V Seismic

design, Design specifications of highway

bridges," Maruzen, Tokyo, Japan.

Kawashima, K, MacRae, G. A., Hoshikuma, J. and

Nagaya, K. (1998). "Residual displacement

response spectra," Journal of Structural

Engineering, 124(5), 513-530, ASCE

Kawashima, K. and Unjoh, S. (1989). "Damping

characteristics of cable-stayed bridges

associated with energy dissipation at movable

support," Structural Engineering and

Earthquake Engineering, Proc. Japan Society of

Civil Engineers, 404/1-11, 123-130.

Kawashima, K., Unjoh, S. and Tsunomoto, M.

(1993). "Estimation of damping ratio of

cable-stayed bridges for seismic design,"

Journal of Structural Engineering, 119(4),

1015-1031, ASCE.

Miranda, E. and Bertero, V. (1994). "Evaluation of

strength reduction factors for earthquake

resistant design," Earthquake Spectra, 10(2),

357-379.

208
Nassar, A. A. and Krawinkler, H. (1991). "Seismic

demands for SDOF and MDOF systems,"

Report No. 95, The John A. Blume Earthquake

Engineering Center, Stanford University,

California, USA.

Newmark, N. M. and Hall, W. J. (1973). "Seismic

design criteria for nuclear reactor facilities,"

Report No. 46, Building Practices for Disaster

Mitigation, National Bureau of Standards, U.S.

Department of Commerce, 209-236.

Press, W.H., Teukolsky, S.A., Vetterling, W.T. and

Flannery, B.P (1996). "Numerical recipes in

Fortran 77," Second Edition, The Art of

Scientific Computing, Cambridge University

Press, 678-683.

Priestley, M. J. N., Seible, F. and Calvi, G. M.

(1996). "Seismic design and retrofit of bridges,"

John Wiley & Sons, New York, USA.

Takeda, T., Sozen, M. A. and Nielsen, N. N.

(1970). "Reinforced concrete response to

simulated earthquake," Journal of Structural

Engineering, 96(12), 2557-2573, ASCE

209
o ^ u.o

.5 0.5

it

Dam

•£3 O

0.4

0.3 Gu

>

0.2

0.1

0, 2 4 6

Ductility Factor

Fig. 1 Definition of Force reduction Factor Fig.2 Equivalent Damping Ratio £,h by Eq. (4)

o

u

§

13uu<

Mto

4>

10

-I 1—I I I Stiff (Type-I)

O Moderate (Type-II) A Soft (Type-nil)

10 100 1000

Epicentral Distance (km)

(a) Soil Condition

u

u

U

CLIO

0.1

.

I CD 8

6.5 -7.0

o 7.0 -7.5

A 7.5 -8.0

-I I I 10 100 1000

Epicentral Distance (km)

(b) Earthquake Magnitude

Fig. 3 Classification of Ground Accelerations in Terms of Soil Conditions and Earthquake

Magnitudes

210
1 2 3

Natural Period (s)

(a)Stiff (Type-I)

1 2 3

Natural Period (s)

(b) Moderate (Type-II)

1 2 3

Natural Period (s)

(c) Soft (Type-Hi)

(1) MT=*

1 2 3

Natural Period (s)

(a) Stiff (Type-I)

2

J

1 2 3

Natural Period (s)

(b) Moderate (Type-II)

24

20

o

-i—i—i—to

ooo

oo

o8

Fig. 4 Force Reduction Factors

1 2 3

Natural Period (s)

(c) Soft (Type-Ill)

(2) Mt =6

211
Average + la r^=/2mt1

Average V V

Average- la

'0 1 2 3

Natural Period (s)

(a) Stiff (Type-I)

12

1 2 3

Natural Period (s)

(b) Moderate (Type-II)

12

2 8

o

0^ 4

0

A

/

IV

0 1 2 3

Natural Period (s)

(a) Stiff (Type-I)

1 2 3

Natural Period (s)

(b) Moderate (Type-II)

'01234

Natural Period (s)

(c) Soft (Type-Ill)

(1) nT =A

Fig. 5 Mean and Mean +/- One Standard

for 70 Ground Motions

0 12 3

Natural Period (s)

(c) Soft (Type-Hi)

(2) nT =6

Deviation of the Force Reduction Factors

212
of

8

6

4

2 F

0

O Stiff (Type-I)

Moderate (Type-II)

a Soft (Type-m)

A

0 12 3

Natural Period (s)

(a) nT =4

1 2 3 4 5

Natural Period (s)

Fig. 8 Idealization of Force Reduction Factors

of

0 1 2 3

Natural Period (s)

(b) fj.j

=6

Fig. 6 Natural Period Dependence of Standard

Deviations of the Force Reduction Factors

10

8

6

4

2

0

1

1 1 1 -e- Stiff (Type-I)

" -€> Moderate(Type-n)

. Soft (Type-IH)

A

A "I

0 2 4 6 8

Target Ductility /V

10

Fig. 7 Target Ductility Factor Dependence of the

Standard Deviations of Force reduction

Factors

Table 1 Parameters a and b and

Regression Coefficients R

(£m,=0.02 and

^= 0.05)

777- a, b Soil Conditions

and

R

Type-I Type-n Type-m

a 1.29 1.12 2.35

2 b 2.77 2.18 1.69

R 0.379 0.701 0.851

a 1.24 0.989 1.52

4 b 2.39 1.62 1.05

R 0.673 0.842 0.886

a 1.34 1.03 1.85

6 b 2.15 1.24 0.821

R 0.717 0.869 0.878

a 1.36 1.20 1.74

8 b 1.67 1.11 0.611

R 0.776 0.899 0.895

213
Stiff (Type-I)

Moderate (Type-II)

Soft (Type-Ill)

2 4 6 8

Target Ductility \ij

(a) Parameter a

Target Ductility ^

(b) Parameter 1/6

10

1

1

) 2 4 6 8 10

0 2 4 6 8

Target Ductility nT

(c) Parameter a + \/b

Fig. 9 Parameters a and a+l/b in Eqs. (20)

0.8

0.6

: 0.4

0.2

0

•a

a"

-Stiff (Type-I)

Moderate (Type-II)

Soft (Type-Ill)

0 2 4 6 8

Target Ductility uT

10

Empirical Model Mean

Mt = 2 O

Mt = 4 n

Mt = 6 A

Mt = 8 X

16

r

12- o

o

8-

of 4.

oi

0

16

12- o

y 8L

1 2 3

Natural Period (s)

(a) Stiff (Type-I)

' %<>e)00oQpu>-0_ a _ ^ _ _o_ _ ^ _

jgggSQgee 8 O 8 O i —1—

1 2 3

Natural Period (s) (b) Moderate (Type-II)

1 2 3

Natural Period (s) (c) Soft (Type-m)

Fig. 11 Application of Eq. (19) to the Mean

Force Reduction Factors Presented in

Fig. 5

Fig. 10 Natural Periods where Force Reduction

Factors Take Values Predicted by the

Equal Energy Assumption (Eq. (15))

214
12

2 8

0

1

1

Qtiff (Time* T\

Moderate (Type-II)

.

Soft (Type-HI)

1 2 3

Natural Period (s)

(a) nT =2

1 2 3

Natural Period (s)

(b) //r

=4

1 2 3

Natural Period (s)

(c) //r =6

1 2 3

Natural Period (s)

(d) jUT =%

Fig. 12 Effect of Soil Condition on the Force

Reduction Factors Predicted by Eq.

(19)

1 2 3

Natural Period (s)

(a) Stiff (Type-I)

1 2 3

Natural Period (s)

(b) Moderate (Type-II)

12 3 4

Natural Period (s)

(c) Soft (Type-HI)

Fig. 13 Force Reduction Factors Corresponding to Means minus One Standard Deviations

215
Table 2 Parameters a and b (%nl =

%EL = °-02) Table 3 Parameters a and b (£nl =

%EL = °-05)

a and

b

Soil Conditions

Type-I Type-n Type-Ill

2 a 0.152 0.225 0.361

b 0.289 1.60 1.12

4 a 0.289 0.348 0.600

b 2.46 1.28 0.902

6 a 0.397 0.432 0.800

b 1.81 1.14 0.768

8 a 0.507 0.513 0.916

b 1.14 1.04 0.632

MT a and

b

Soil Conditions

Type-I Type-II Type-in

2 a 0.226 0.344 0.521

b 4.14 1.94 1.34

4 a 0.778 0.572 0.976

b 3.50 1.35 0.994

6 a 0.981 0.725 1.23

b 2.93 1.15 0.757

8 a 1.23 0.807 1.28

b 2.57 0.983 0.569

(/J

^

'

C/5

03

0

0

0

Snl=Sel=0.02

Snl = 0.02, 4el=005

Snivel =0-5

2 4 6 8

Target Ductility Mr

(a) Stiff (Type-I)

2 4 6

Target Ductility

(b) Moderate (Type-II)

2 4 6 8

Target Ductility ^

(c) Soft (Type-IH)

(1) Parameter a

2 4 6 8

Target Ductility

(a) Stiff (Type-I)

2 4 6 8

Target Ductility Uj

(b) Moderate (Type-II)

2 4 6 8

Target Ductility Mt

(c) Soft (Type-m)

(2) Parameter a+l/b

10

10

10

Fig. 14 Dependence of Parameters a and a+ l/b on the Assumption of Damping Ratios

216
15 -

£ io h

Present Study Means

SNL

= f -

SEL fl CO U

= 0.02, ^=0.05

^NL

=

" 4el = 0.5 A

....

,

1 1 i i i

o

1

i i i 0 12 3

Natural Period (s)

(a)Stiff (Type-I)

i r

1 2 3

Natural Period (s)

(b) Moderate (Type-II)

"i 1 r

o o a -iV -u.-B1 2 3

Natural Period (s)

(c) Soft (Type-Ill)

(1) Mr**

20

o

20

15 -

-:=> o

1 1 1 —1 1— To

o ° 0

^

A

'

A

1 1 1 1 1 1

1 2 3

Natural Period (s)

(a) Stiff (Type-I)

~i 1 1 r

1 2 3

Natural Period (s)

(b) Moderate (Type-II)

~t r

1 2 3

Natural Period (s)

(c) Soft (Type-ffl)

(2) //r

=8

Fig. 15 Dependence of Force Reduction Factors on the Assumption of Damping Ratios

217
Present Study Nassaretal.

Miranda et al. Means

16

r

12

Ih 2

u

8

4

0

0

16

f

12

Ino

«4H 8

4

0

0

16

f

12

o

o

J? 8 (4-1

4

0

0

1 2 3

Natural Period (s)

(a) Stiff (Type-I)

NI»bii.._Ll_._l_i... ti .M^l

1 2 3 Natural Period (s)

(b) Moderate (Type-II)

1 2 3

Natural Period (s)

(c) Soft (Type-Ill)

16r

12

IhOo

8

<*H Pf

4-

0-

0

16

12

Iho*Ho

8

<4H of

4

0

1 2 3

Natural Period (s)

(a) Stiff (Type-I)

1 2 .3

Natural Period (s) (b) Moderate (Type-II)

1 2 3

Natural Period (s)

(c) Soft (Type-HI)

(1) ^r=4 (2) Mt=*

Fig. 16 Comparison with Previous Studies

218
Bolu Viaduct-1 Subjected to Near-Fault Ground Motion

by

Sunwoo Park1 , Hamid Ghasemi2

, Jerry Shen1

, and Phillip Yen2

ABSTRACT

The performance of seismically-isolated Bolu

Viaduct-1 in Turkey subjected to a simulated

near-fault ground motion during the 1999 Duzce

Earthquake was evaluated through nonlinear

finite element analysis. The ground motion is characterized by a large residual movement of

the ground across the fault rupture that crosses

the viaduct. Analysis indicates that the ground

motion induces response that exceeds the design

capacities of the seismic isolation systems thus

resulting in substantial damage to the bearings

and energy dissipation units, which is consistent

with post-earthquake field observation. The

analysis also indicates that shear keys, both

longitudinal and transverse, play a critical role in

preventing the superstructure collapse.

KEY WORDS: seismic isolation, damping,

Viaduct 1, Duzce earthquake, near-fault

1.0 INTRODUCTION

Developments in seismic isolation and energy

dissipation devices have permitted considerable

advances in seismic protection of highway

bridges. However, existing seismic protection

strategies are largely based on design ground

motions that do not contain near-fault features.

Experience with a number of recent earthquakes

has pointed to the need for accounting for the

effects of near-fault ground motions in the

seismic design of highway bridges. In this paper,

the effects of a near-fault earthquake ground

motion on the performance of a seismically

isolated bridge are studied through numerical

analysis of Bolu Viaduct 1 subjected to the 1999

Duzce Earthquake.

Bolu Viaduct 1 is located in north central

Turkey and is part of the Trans-European

Motorway (TEM) running from Ankara to Europe, see Figs 1 and 2. The 2. 3 -km viaduct,

with its 59 dual spans, was approximately 95%

complete at the time of the earthquake. The

superstructure consists of seven lines of simplysupported, prestressed-concrete box girders

seated on sliding bearings with stainless

steel/PTFE slider interfaces. The deck slab is monolithic over 10-span segments and each

segment is 392 m long and supported by 11

piers. An energy dissipation system of yieldingsteel type is also installed on each pier cap to form, together with the sliding bearings, a seismic isolation system for the viaduct.

2.0 NUMERICAL ANALYSIS

Numerical simulation of the response of a

typical ten-span segment of the viaduct

subjected to a recorded and a simulated

earthquake ground motion is conducted via

nonlinear time-history analysis. The viaduct's

superstructure and piers are modeled using 3-D

beam elements, and the foundation conditions at each pier base are modeled by elastic

translational and rotational springs. Special

nonlinear link elements are used to model

energy dissipation units (EDUs), sliding (or

friction pot) bearings, and shear keys, see Fig. 3. The schematic illustration of the constitutive

behavior of the nonlinear link elements used in

the analysis are presented in Fig. 4. The forcedisplacement hysteretic behavior of the EDU

resembles that of a lead-rubber bearing or

friction pendulum system, and behavior of the

sliding bearing is dictated by the coefficient of

1 LENDIS Corporation @ Federal Highway Administration, Office of Infrastructure R&D, 6300

Georgetown Pike, McLean, VA 22101, USA

2 Federal Highway Administration, Office of Infrastructure R&D, 6300 Georgetown Pike, McLean, VA

22101, USA

219
friction of the sliding surfaces. Seven bearings,

each between a girder and pier cap, are modeled

by a single bearing element, and multiple

transverse shear keys are also modeled by a

single gap/hook element, situated between the

superstructure and pier. A pedestal for an EDU

plays the role of a longitudinal shear key. The

displacement capacity of the sliding bearings is 210 mm and the displacement capacities of the

EDUs, longitudinal and transverse shear keys

are all 480 mm, which presents apparent

inconsistency.

A ground motion recorded at the Bolu station in

the 1999 Duzce earthquake and a simulated

near-fault ground motion are used as the input in

the analysis. Figure 5 shows the acceleration

histories and spectra of the two horizontal

components of the Bolu ground motion. The

design spectra shown is based on the 1999

AASHTO Guide Specifications for Seismic

Isolation Design (AASHTO, 1999). An

acceleration coefficient, A=0.4, and Soil Type II are assumed. The acceleration histories and

spectra of the fault-normal and fault-parallel

components of the simulated near-fault ground

motion at the site of Viaduct 1 are shown in Fig.

6, and the corresponding ground displacement

histories are presented in Fig. 7. It is to be noted

that the effect of surface fault rupture that crosses the 10-span segment is modeled by

specifying a different ground motion to each

side of the segment divided by the rupture

crossing as indicated in Fig. 8. The surface

rupture is oriented at an approximate angle of

25° relative to the axis of the viaduct segment

analyzed. The left-hand side of the segment in

Fig. 8 is subjected to a motion defined by faultnormal and fault-parallel 1 components shown

in Fig. 6, and the right-hand side is subjected to a motion defined by fault-normal and faultparallel 2 components.

The two fault-parallel components are created

by superimposing a pulse motion ("Type-A" by

Makris and Chang, 2000) on the E-W

component of the Bolu record, and the faultnormal component by superimposing a pulse

motion ("Type-B" by Makris and Chang, 2000)

on the N-S component of the Bolu record. The

two fault-parallel components contain a pulse

motion that induces a residual displacement of

0.75 m in opposite directions (representing

"fling" effect), thus resulting in a relative

displacement of 1 .5 m across the surface rupture, see Fig. 7. The ground motions applied to either

side of the segment have the same fault-normal

component due to kinematic continuity across

the fault rupture, and the magnitude of the faultnormal component is estimated from its conjugate relationship to the fault-parallel

component in a related problem (Anderson and

Luco, 1983). The simulated near-fault ground

motion uses the Bolu record as the background

far-field motion and is consistent with the field

observation of an approximate relative ground

movement of 1 .5 m across the surface rupture.

3.0 RESULTS OF ANALYSIS

Analysis has focused on the displacement of

isolation/damping elements. The bearing

displacement at the center pier of the 10-span

segment subjected to the Bolu ground motion is shown in Fig. 9. It is to be noted that the

displacement of bearing (or EDU) represents the

relative displacement between the superstructure

and pier. The time histories of the longitudinal

and transverse components are shown in Figs.

9(a) and (b), respectively, and the displacement

path on the bearing surface in Fig. 9(c). The

displacement capacities of sliding bearings and

shear keys are indicated in broken lines in Fig.

9(c). It can be seen that the bearing's

displacement capacity is exceeded at an early

stage of the movement. It is believed that, due to

this exceedance, EDUs are distorted and failed

before they can be functioned as designed. Postearthquake field investigations have revealed

that the majority of the EDUs were broken for

the 10-span segment which was crossed by the

fault rupture. In our analysis, different levels of

EDU survival were considered, and the results

shown in Figs. 9 to 11 are all based on the

assumption that the entire EDUs are failed at the

beginning. Figure 9 indicates that, while the

longitudinal shear key was briefly engaged, the

transverse shear key has never been engaged

under the Bolu ground motion. Field inspections

revealed that transverse as well as longitudinal

shear keys of Viaduct 1 were repeatedly engaged

during the earthquake. It is believed that the

220
Bolu ground motion does not induce the

observed damage to the viaduct, suggesting that a much stronger near-fault ground motion must

have hit the Viaduct 1 site in the 1999 Duzce

earthquake.

Figures 10 and 11 respectively show the bearing

displacements at the pier immediately left (PI in

Fig. 8) and immediately right (P2 in Fig. 8) of

the fault rupture under the simulated near-fault

ground motion. The displacement capacity of the

sliding bearings is exceeded during the first major cycle of movement, and then both

longitudinal and transverse shear keys are engaged. Analysis indicates that the longitudinal

shear keys at both piers remain engaged after the

event due to the large relative ground movement

across the fault rupture, and that the transverse

shear key at PI remain engaged following

several interactions between the girder and shear

key. The displacement demand due to the static

ground movement (1.5 m) is accommodated

partly by bearings (until the longitudinal shear

key is engaged) and partly by the pier. It is interesting to note that the shear keys, both

longitudinal and transverse, are engaged only on

one side throughout the motion, which is consistent with findings from field inspection of

the shear keys.

Finally, post-earthquake field reconnaissance

survey has revealed that the majority of the

stainless steel interface plates had distinct

scoring that resembled the number "9" as shown

in Fig. 12(a) (Ghasemi et al., 2000). The

calculated displacement path as shown in Fig.

12(b) is fairly consistent with the field

observation. The initial movement consists of a half cycle of displacement with an amplitude of

approximately 130 mm followed by movement

in the opposite direction that exceeds the

bearing's displacement capacity. Due to this

exceedance at an early stage of the ground

motion, subsequent large movement of the

girders (which are already dropped off the

bearing plates) on pier-cap table is believed to have caused damage to sliding bearings

(including ejection of bearing plates) and EDUs

and, as a consequence, the entire seismic

isolation/damping systems did not function as designed. However, from a technical point of

view, the viaduct performed satisfactorily

largely due to shear keys (longitudinal and

transverse) that prevented collapse of the

superstructure segments.

4.0 CONCLUSIONS

Analysis of the seismically-isolated Viaduct 1 subjected to the ground motion recorded at the

Bolu station resulted in displacements exceeding

the capacity of the isolation bearings but not

large enough to engage shear keys and to cause

significant damage to the viaduct as was

observed in the post-earthquake investigation.

However, analysis with a simulated near-fault

ground motion that accounts for relative static

ground movement across the surface rupture

resulted in displacements far exceeding the

capacities of the isolation bearings and EDUs

and resulted in engagement of the shear keys,

both longitudinal and transverse. The calculated

bearing displacement paths are consistent with

the "9"-shaped scoring trace observed on the

surface of the bearing plates at the site. The

close proximity of the fault rupture to the

viaduct caused significant superstructure

movement relative to the piers resulting in severe damage to bearings, EDUs, and shear

keys. The simulation, however, indicates that the

shear keys played a critical role in preventing

superstructure segments from dropping off the

pier caps thus preventing a collapse of the

viaduct during the earthquake.

5.0 REFERENCES

AASHTO (1999). Guide Specifications for

Seismic Isolation Design. American Association

of State Highway and Transportation Officials,

Washington, DC.

Anderson, J. G. and Luco, J. E. (1983).

Parametric study of near-field ground motion for a strike-slip dislocation model. Bulletin of the

Seismological Society ofAmerica, 73(1), 23-43.

Makris, N. and Chang, S.-P. (2000). Effect of

viscous, viscoplastic, and friction damping on

the response of seismic isolated structures. Earthquake Engineering and Structural

Dynamics, 29, 85-107.

221
Investigation of the Structures on the TEM.

Ghasemi, H., Cooper, J. D., Imbsen, R., Piskin, Publication No. FHWA-RD-00-146, Federal

H., Inal, F., and Tiras, A. (2000). The November Highway Administration, U. S. Department of

1999 Duzce Earthquake: Post-Earthquake Transportation, Washington, D.C.

20 mt9i

ZONdtlLOAK ''

Fig.l. General area map.

Fig.2. General view of Bolu Viaduct 1

222
I- 10 @ 39m

XL

3-D Beam «

»

Elements

-o——o—oNonlinear Link Elements

modeling

-Energy Dissipation Unit (EDU)

-Friction Pot Bearings, and

-Shear Keys

Foundation Support

modeled by Springs

Fig. 3. Finite element discretization of a 10-span segment of Viaduct 1

Force Force

Displacement

(a) (b)

Displacement

Fo

-D

ce

+D

(c)

Displacement

Fig. 4. Schematic illustration of constitutive models for (a) EDUs, (b) friction bearings, and (c) shear keys.

223
3 0.5 c

o

*5

<0 0

i_

0>

0)

cc -0.5 <

-1

•—

' 0.5 C

g

ra o

a>S

4

-0.5

2.5

3 2

-

0

10

10

Bolu (N-S) PGA = 0.728 g

20 30 40 50

Bolu (E-W) PGA = 0.822 g

20 30 Time (s)40 50

AASHTO 1999

Bolu (N-S)

Bolu (E-W)

2 3

Period (s)

Fig. 5. Acceleration histories and spectra of the ground motion recorded at the Bolu station.

224
0

20 30 40

Fault-Parallel 2 pga = 0.742 g

i i i i i

10 20 30 Time (s)

40

2 3

Period (s)

50

50

Fig. 6. Acceleration histories and spectra of the simulated near-fault ground motion.

225
E

*-< c

<DE

a>o

_rea

wQ

100

80

60

40

20

0

-20

- Fault-Normal

10 20 30 Time (s)40 50

Fig. 7. Displacement histories of the simulated near-fault ground motion.

PI P2

Nor

KthEast

0.75 m

^25°

0.75 m

Surface Fault Rupture

Fig. 8. Orientation of the surface fault rupture and direction of the residual ground movement.

226
E

a.

00

(0

c

3

60

30

0

O) -30

c

o

-60

\

(a)

1 1

V l|\ a fKJ^'^r-—

• i i

60

E

cL

CO b

(D

s

00c

(0

30 -

-30

-60

E

o

Q_

0)

(1)

</) i_

0

>

c

CD

60

30

0 -

30 -

-60

10 15 20 25 30

(b)

\

1 1

i i i

Bearing

Disp. Capacity

(R=21cm)

Shear-Key

Disp. Capacity

(48cm x 48cm)

-60 -30 0 30 60

Longitudinal Disp. (cm)

Fig. 9. Bearing displacement at the center pier under the Bolu ground motion: (a) longitudinal displacement

history, (b) transverse displacement history, and (c) displacement path on the bearing surface.

227
E

cL

w

COc

3

60

30

0

D) -30

c

o

-60

1

(a)

A/

L

0

60

E

30 - cL

(0

b

0

-

CD>

c -30 - 2

i-

-60

E

CL

CO

CD

CO

i_

CD

>

CO

c

03

60

30 -

0 -

30 -

10 15 20

10 15 20

Time (s)

25 30

25 30

-60

-60 -30 0 30 60

Longitudinal Disp.(cm)

Fig. 10. Bearing displacement at Pier PI (see Fig. 8) under the simulated near-fault ground motion:

(a) longitudinal displacement history, (b) transverse displacement history, and (c) displacement path.

228
E

cL

co

COc

3

60

30

0

en -30

c

o

-60

sj-i

(a)

E

CL

CO b

CD

S

COc

CO

60

E

o

Q_

0

c/)

30

0 - 0

1-30

03

-60

5 10 15 20 25 30

5 10 15 20

Time (s)

25 30

-60 -30 0 30 60

Longitudinal Disp.(cm)

Fig. 11. Bearing displacement at Pier P2 (see Fig. 8) under the simulated near-fault ground motion:

(a) longitudinal displacement history, (b) transverse displacement history, and (c) displacement path.

229
30

E 20

o

co

1 u

0

CO

CD

>

CO

c

2

0 -

-10

-20 -

-30

(b)

-30 -20 -10 0 10 20 30

Longitudinal Disp.(cm)

Fig. 12. Displacement path on the bearing surface during the initial stage of movement: (a) observed scoring

trace on the bearing plate and (b) calculated displacement path from finite element analysis.

230
2002 Seismic Design Specifications for Highway Bridges

by

Shigeki Unjoh1

, Sho-ichi Nakatani2, Kei-ichi Tamura3

Jiro Fukui4 and Jun-ichi Hoshikuma5

ABSTRACT

This paper presents the revised Seismic Design

Specifications of Highway Bridges, Japan Road

Association (JRA), in 2002 [1]. The revised

specifications are based on the Performance-based

Design Code Concept and the improved

knowledges on the seismic design methods for

highway bridges are also included. According to

the Performance-based Design Code Concept, the

design requirements are clearly specified and the

existing detailed design methods including the

analytical methods and the allowable limits are

specified as verification methods and/or the

examples of acceptable solutions. The designers can modify or select other verification methods

and/or acceptable solutions with appropriate

verifications. It is expected that new ideas on the

materials and structures will be employed much

easier than before. In this paper, the revised major

points in the 2002 JRA Seismic Design

Specifications are briefly described including the

performance-based design code concept, seismic

performance levels, the seismic design force, and

the ductility evaluation methods for foundations

and superstructures.

KEYWORDS: Seismic Design

JRA Design Specifications

Highway Bridges

Performence-based Design Code

1. INTRODUCTION

Seismic design methods for highway bridges in

Japan has been developed and improved based on

the lessons learned from the various past bitter

experiences after the Great Kanto Earthquake

(M7.9) in 1923. By introducing the various

provisions for preventing serious damage such as

the design method against soil liquefaction, design

detailing including the unseating prevention

devices, a number of highway bridges which

suffered complete collapse of superstructures was

only a few in the recent past earthquakes.

However, the Hyogo-ken-Nanbu Earthquake of

January 17, 1995, caused destructive damage to highway bridges. Collapse and nearly collapse of

superstructures occurred at 9 sites, and other

destructive damage occurred at 16 sites [2, 3]. The

earthquake revealed that there are a number of

critical issues to be revised in the seismic design

and seismic strengthening of bridges. Based on

the lessons learned from the Hyogo-ken-Nanbu

1 Leader, Earthquake Engineering Team, Earthquake Disaster Prevention Research Group, Public Works

Research Institute (PWRI), Tsukuba-shi, Ibaraki-ken, 305-8516 Japan

2 Head, Bridge Division, Road Department, National Institute for Land and Infrastructure Management

(NILIM), Ministry of Land, Infrastructure and Transport, Tsukuba-shi, Ibaraki-ken, 305-0804 Japan

3 Leader, Ground Vibration Team, Earthquake Disaster Prevention Research Group, PWRI

4 Leader, Foundation Engineering Team, Structures Research Group, PWRI

5 Senior Research Engineer, Earthquake Engineering Team, Earthquake Disaster Prevention Research

Group, PWRI

231
Earthquake, the design specifications for highway

bridges were significantly revised in 1996 [3, 4, 5].

The intensive earthquake motion with a short

distance from the inland earthquakes with

Magnitude 7 class as the Hyogo-ken-Nanbu

Earthquake has been considered in the design.

In the Japan Road Association (JRA), the "Bridge

Committee" (Chairman: Dr. Syo-ichi SAEKI,

Director of Public Works Research Center) has

worked on the revision of the design

specifications of highway bridges. The target

point of the revision was to be based on the

performance-based design code concept and to enhance the durability of bridge structures for a long-term use, as well as the inclusion of the

improved knowledges on the bridge design and

construction methods. The new Design

Specifications of Highway Bridges was issued by

the Ministry of Land, Infrastructure and Transport

on December 27, 2001. The JRA has released it

with the commentary in March 2002. This paper

summarizes the new JRA Design Specifications of

Highway Bridges, Part V: Seismic Design, issued

in March 2002.

2. MAJOR REVISION OF JRA SEISMIC

DESIGN SPECIFICATIONS

The major revision point is to be based on the

performance-based design code concept.

According to the performance-based design code

concept, the code structure, in which both the

design requirements and the existing detailed

design methods are clearly separated and specified,

is employed. And the improved knowledges on

the seismic design methods are also included.

The major revisions of the Part V: Seismic Design

are as follows:

(1) Based on the performance-based design code

concept, principle requirements on the seismic

performance of highway bridges, determination

concept of design earthquake ground motion and

principle to verify the seismic performance are

clearly specified.

(2) Two earthquake level design concept is used

and the design earthquake ground motion with

high probability to occur and the design

earthquake ground motion with high intensity and

low probability to occur is employed as the same

as 1996 JRA Specifications. The ground motions

are named as Level 1 Earthquake and Level 2

Earthquake, respectively.

(3) Verification methods of seismic performance

are rearranged as "Static Analysis" and "Dynamic

Analysis." The selection of two design methods is

clearly shown. The applicability of the dynamic

analysis is much widened and the detailed

verification method for the dynamic analysis is

specified.

(4) The evaluation method of dynamic earth

pressure for the Level 2 Earthquake design is

introduced. This is the based on the modified

Mononobe-Okabe earth pressure theory. The

evaluation method of the dynamic water pressure

for the Level 2 Earthquake design is also

introduced.

(5) The verification method of the seismic

performance of abutment foundations on the

liquefiable ground is newly introduced.

(6) The evaluation method of the

force-displacement relation models for steel

columns with/without infilled concrete is improved.

(7) The verification method of the seismic

performance for steel and concrete superstructures

are newly introduced.

(8) The evaluation methods of the strength for

bearing supports are improved.

(9) References on the back data of the design

methods and related information are added at the

end of the specifications.

3. PERFORMANCE-BASED DESIGN

SPECIFICATIONS

The JRA Design Specifications has been revised

232
based on the Performance-based design code

concept for the purpose to respond the

international harmonization of design codes and

the flexible employment of new structures and

new construction methods. The

performance-based design code concept is that the

necessary performance requirements and the

verification policies are clearly specified. The JRA

specifications are employed the style to specify

both the requirements and the acceptable solutions

including the detailed performance verification

methods which are based on the existing design

specifications including the design methods and

the design details. For example, the analysis

method to evaluate the response against the loads

is placed as one of the verification methods or

acceptable solutions. Therefore, designer can

propose the new ideas or select other design

methods with the necessary verification.

The most important issue of the

performance-based design code concept is the

clear specifications of the requirements, which the

designers are not allowed to select other methods,

and the acceptable solutions, which the designers can select other methods with the necessary

verification. In the JRA Specifications, they are

clearly specified including the detailed

expressions. In future, the acceptable solutions

will be increased and widened with the increase of

the verification of new ideas on the materials,

structures and constructions methods.

The code structure of the Part V: Seismic Deign is

as shown in Fig. 1. The static and dynamic

verification methods of the seismic performance

as well as the evaluation methods of the strength

and ductility capacity of the bridge members are

placed as the verification methods and the

acceptable solutions, which can be modified by

the designers with the necessary verifications.

4. BASIC PRINCIPLES OF SEISMIC DESIGN

Table 1 shows the performance matrix including

the design earthquake ground motion and the

Seismic Performance Level (SPL) provided in the

revised JRA Seismic Design Specifications in

2002. There is no revision on this basic principle

from the 1996 Version.

The two level ground motion as the moderate

ground motions induced in the earthquakes with

high probability to occur (Level 1 Earthquake)

and the intensive ground motions induced in the

earthquakes with low probability to occur (Level

2 Earthquake).

The Level 1 Earthquake provides the ground

motions induced by the moderate earthquakes and

the ground motion considered in the elastic design

method in the past for a long time is employed.

For the Level 2 Earthquake, two types of ground

motions are considered. The first is the ground

motions which is induced in the

interplate-typeearthquakes with the magnitude of

around 8. The ground motion at Tokyo in the 1923

Kanto Earthquake is a typical target of this type of

ground motion. The second is the ground motion

developed in earthquakes with magnitude of

around 7 at very short distance. The ground

motion at Kobe during the Hyogo-ken-Nanbu

Earthquake is a typical target of this type of

ground motion. The first and the second ground

motions are named as Type-I and Type-II ground

motions, respectively. The recurrence period of

the Type-II ground motion may be longer than

that of the Type-I ground motion, although the

estimation is very difficult.

In the 2002 revision, the design ground motions

are named as Level 1 Earthquake and Level 2

Earthquake. One more important revision on the

design earthquake ground motion is that the

site-specific design ground motions shall be

considered if the ground motion can be

appropriately estimated based on the information

on the earthquake including past history and the

233
location and detailed condition of the active faults,

ground conditions including the condition from

the faults to the construction sites. To determine

the site-specific design ground motion, it is

required to have the necessary and accurate

informations on the earthquake ground motions

and ground conditions as well as the verified

evaluation methodology of the fault-induced

ground motions. However, the area to get such

detailed informations in Japan is very limited so

far. Therefore, the continuous investigation and

research on this issue as well as the reflection on

the practical design of highway bridges is

expected.

5. GROUND MOTION AND SESIMIC

PERFORMANCE LEVEL

The seismic design of bridges is according to the

performance matrix as shown in Table 1. The

bridges are categorized into two groups depending

on their importances; standard bridges (Type-A

bridges) and important bridges (Type-B bridges).

Seismic Performance Level (SPL) depends on the

importance of bridges. For the moderate ground

motions induced in the earthquakes with high

probability to occur, both A and B bridges shall

behave in an elastic manner without essential

structural damage (Seismic Performance Level

(SPL): 1). For the extreme ground motions

induced in the earthquakes with low probability to

occur, the Type-A bridges shall prevent critical

failure (SPL: 3), while the Type-B bridges shall

perform with limited damage (SPL: 2)

.

The SPLs 1 to 3 are based on the viewpoints of

"Safety," "Functionability," "Repairability" during

and after the earthquakes. Table 2 shows the basic

concept of these three viewpoints of the SPL.

6. VERIFICATION METHODS OF SEISMIC

PERFORMANCE

6.1 Seismic Performance Level and Limit States

As mentioned in the above, the seismic

performance is specified clearly. It is necessary to determine and select the limit states of highway

bridges corresponding to these seismic

performance levels to attain the necessary

performance in the design procedure of highway

bridges.

In the 2002 revision, the determination principles

of the limit state to attain the necessary seismic

performance are clearly specified. For example,

the basic principles to determine the limit state for

SPL 2 is: 1) the plastic hinges shall be developed

at the expected portions and the capacity of plastic

hinges shall be determined so that the damaged

members can be repaired relatively easily and

quickly without replacement of main members, 2)

the plastic hinges shall be developed at the

portions with appropriate energy absorption and

with high repairability, 3) considering the

structural conditions, the members with plastic

hinges shall be combined appropriately and the

limit states of members with plastic hinges shall

be determined appropriately. Based on the basic

concept, the combinations of members with

plastic hinges and the limit states of members for

ordinary bridge structures are shown in the

commentary.

6.2 Verification Methods of Seismic

Performance

It is the fundamental policy of the verification of

seismic performance that the response of the

bridge structures against design earthquake

ground motions does not exceed the determined

limit states.

Table 3 shows the applicable verification methods

of seismic performance used. In the seismic

design of highway bridges, it is important to

increase the strength and the ductility capacity to

appropriately resist the intensive earthquakes. The

verification methods are based on the static

234
analysis and dynamic analysis. In the 1996 design

specifications, the lateral force coefficient

methods with elastic design, ductility design

methods and dynamic analysis were specified and

these design methods had to be selected based on

the structural conditions of bridges. The basic

concept is the same as 1996 one but the

verification methods are rearranged to the

verification methods based on static and dynamic

analyses.

The static verification methods including the

lateral force design method and the ductility

design method are applied for the bridges with

simple behavior with predominant single mode

during the earthquakes. The dynamic verification

method is applied for the bridges with

complicated behavior, in such case the

applicability of the static verification methods is

restricted. In the 1996 design specifications, for

the bridges with complicated behavior both the

static and dynamic analyses had to be applied and

satisfied. In the 2002 one, the applicability of the

dynamic analysis is widened and the dynamic

verification method is expected to be used mainly

with appropriate design consideration.

6.3 Major Revision Points of the Verification

Methods of Seismic Performance

6.3.1 Verification Methods of

Abutment-Foundation on Liquefiable Ground

against Level 2 Earthquake

In the 1996 design specifications, the performance

of the abutment-foundations was not verified in

detail. This is because 1) the serious damages to abutment-foundations were not found in the past

earthquakes when the soil liquefaction was not

developed, 2) abutment-foundation is affected by

the backfill soils during earthquakes and the effect

of the inertia force of abutment itself is relatively

small comparing with the pier-foundations, 3)

since abutments generally resist against back-fill

earth pressure, the abutment-foundations tend to develop displacement to the direction of the earth

pressure that is to the center of bridges, then it is generally low probability to have the unseating of

superstructures.

On the other hand, recently, the dynamic earth

pressure against Level 2 Earthquake based on the

modified Mononobe-Okabe theory has been

proposed and the behavior of the

abutment-foundations can be evaluated during the

Level 2 earthquakes. Based on investigations

using the modified Mononobe-Okabe theory, it is shown that the abutment-foundations designed

according to the Level 1 Earthquake generally

satisfy the performance requirement during the

Level 2 Earthquake. Therefore, based on these

results, the performance of the

abutment-foundations only on the liquefiable

ground shall be verified in order to give the

necessary strength to the foundations and to limit

the excessive displacement even if the nonlinear

behavior is expected in the abutment-foundations.

6.3.2 Verification Methods of Strength and

Ductility of Steel Column

In the 1996 design specifications, the concrete

infilled steel columns was designed according to

the static ductility design methods using the

response evaluation based on the energy equal

theory. The force-displacement relation was based

on the experimental data of steel columns. On the

other hand, steel columns without infilled

concrete was designed based on the dynamic

analysis because the applicability of the static

response evaluation was not verified.

In the 2002 design specifications, new and more

appropriate force-displacement relation models

for steel columns with and without infilled

concrete are proposed based on the experimental

data of steel columns which has been made before

and after the last 1996 revision. Using these new

235
models, the seismic performance is verified based

on the dynamic analysis.

6.3.3 Verification Methods of Strength and

Ductility of Superstructure

Generally, the seismic design of superstructures is

not critical except the portion around the bearing

supports which are the connection between

superstructure and substructures. However, the

seismic design sometimes becomes critical in the

design of rigid frame bridges and arch bridges in

the longitudinal direction, and in the design of

bridges with relatively long spans to the bridge

width in the transverse direction.

The nonlinear behavior of superstructures against

cyclic loading is investigated in the recent

research. Therefore, the verification method of the

limited nonlinear performance for the

superstructures is newly specified with the

assumption of energy absorption at the plastic

hinges in the columns.

7. CONCLUDING REMARKS

This paper presented an outline of the revised JRA

Seismic Design Specifications of Highway

Bridges issued in 2002. Based on the lessons

learned from the Hyogo-ken-Nanbu Earthquake in

1995, the "Part V: Seismic Design" of the "JRA

Design Specifications of Highway Bridges" was

totally revised in 1996, and the design procedure

moved from the traditional Seismic Coefficient

Method to the Ductility Design Method. Major

point of the revision was the introduction of

explicit two-level seismic design methods. In the

2002 revision, the target point of the revision is to be based on the performance-based design code

concept and to enhance the durability of bridge

structures for a long-term use, as well as the

inclusion of the improved knowledgs on the

bridge design and construction methods. It is expected to have the circumstances to employ the

new ideas on the materials, structures and

constructions methods to construct safer, more

durable and more cost-effective bridges in the

future.

Acknowledgments

Drafting of the revised version of the "Part V

Seismic Design" of the "2002 JRA Design

Specifications of Highway Bridges" was

conducted at the "Sub-committee for Seismic

Design of Highway Bridges" and was approved

by the Bridge Committee, Japan Road Association.

The authors served as members of bridge

committee and executive member of the

sub-committees, respectively. The authors thank

all members of the Bridge Committee and the

Sub-Committees.

REFERENCES:

1) Japan Road Association: Design Specifications

of Highway Bridges, Part I Common Part, Part II

Steel Bridges, Part III Concrete Bridges, Part IV

Foundations, and Part V Seismic Design, 2002

2) Ministry of Construction: Report on the

Damage of Highway Bridges by the Hyogo-ken

Nanbu Earthquake, Committee for Investigation

on the Damage of Highway Bridges Caused by

the Hyogo-ken Nanbu Earthquake, 1995

3) Japan Road Association: Design Specifications

of Highway Bridges, Part I Common Part, Part II

Steel Bridges, Part III Concrete Bridges, Part rv

Foundations, and Part V Seismic Design, 1996

4) Kawashima, K.: Impact of Hanshin/Awaji

Earthquake on Seismic Design and Seismic

Strengthening of Highway Bridges, Report No.

TIT/EERG 95-2, Tokyo Institute of Technology,

1995

5) Kazuhiko Kawashima, et. al.: 1996 Design

Specifications for Highway Bridges, 29th UJNR

Joint Panel Meeting, May 1996

236
APPENDIX

2002 JRA Design Specifications for HighwayBridges, Part V: Seismic Design

Table of Contents

3.

General

1 . 1 Scope

1 .2 Definition of Terms

Basic Principle of Seismic design

2.1 Policy of Seismic Design

2.2 Basic Principles of Seismic Design

2.3 Definition of Importance of Highway Bridges

Loads to be considered in Seismic Design

9. Verification of Seismic Performance of Seismically-Isolated

Bridges

9.1 General

9.2 Verification of Seismic Performance

9.3 Design Model of Isolators

9.4 Basic Requirements for Isolators

9.5 Other Structures for reducing Effects of Earthquake

10. Reinforced Concrete Columns

3.1 Loads to be considered in Seismic Design and the 10.1 Combinations

3.2 Effects of Earthquake

4. Design Earthquake Ground Motion

4.1 General

4.2 Level 1 Earthquake

4.3 Level 2 Earthquake

4.4 Modification Factor of Zone

4.5 Category of Ground Conditions

4.6 Ground Level to be considered in Seismic Design

5. Verification of Seismic Performance

5.1 General

5.2 Limit State for Seismic Performance Level (SPL) 1

5.3 Limit State for Seismic Performance Level (SPL) 2

5.4 Limit State for Seismic Performance Level (SPL) 3

5.5 Basic Principles to verify Seismic Performances

5.6 Countermeasures for Preventing Unseating

Superstucture

General

10.2 Evaluation of Damage Mode, Strength and Ductility

Capacity

1 0.3 Evaluation of Strength and Displacement

10.4 Stress-Strain Relation of Concrete

10.5 Shear Strength

10.6 Design Details to increase Ductility Performance

10.7 Termination of Longitudinal Reinforcement at Mid-Heigth

10.8 Strength and Ductility Capacity of Reinforced Concrete

Two Column Bents

10.9 Effect of Eccentric Loading of Superstructure

1 1 . Steel Columns

11.1 General

1 1 .2 Verification of Seismic Performance of Steel Columns by

Dynamic Analysis

of 1 1 .3 Design Details

1 1 .4 Design of Anchor of Columns

Verification Methods of Seismic Performance based on 12. Pier-Foundation

Static Analysis

6.1 General

6.2 Evaluation of Loads for Static Analysis

6.3 Verification of Seismic Performance for Level

Earthquake

6.4 Verification of Seismic Performance for Level

Earthquake

12.1 General

12.2 Strength, Reaction of Soils and Displacement of

Pier-Foundation

1 12.3 Yield of Pier-Foundation

1 2.4 Ductility Response of Pier-Foundations when Foundation

2 behaves in Nonlinear Region

1 2.5 Ductility and Displacement Capacity of Pier-Foundation

7. Verification Methods of Seismic Performance based on 12.6 Design of Members of Pier-Foundation

Dynamic Analysis

7.1 General

7.2 Earthquake Ground Motion for Dynamic Analysis

7.3 Analytical Model and Analysis Methods

7.4 Verification of Seismic Performance

Effect of Unstable Ground Condition during Earthquake

8.1 General

8.2 Design Soil Coefficient of Soft Soil and Liquefiable 14.1 Sandy Soils 14.2

13. Abutment-Foundation on Liquefiable Ground

13.1 General

Design Seismic Force for Abutment-Foundation

Ductility Response of Abutment-Foundation

Ductility Capacity of Abutment-Foundation

Design of Members of Abutment-Foundation

Superstructure and Deck End Structure General

Steel Superstructure

13.2

13.3

13.4

13.5

14.

8.3 Effect of Lateral Spreading induced

Liquefaction and the Design Methods

by Soil 14.3 Concrete Superstructure

14.4 Deck End Structure

237
15. Bearing Support

15.1 General

15.2 Design Seismic Force for Bearing Support

1 5.3 Design of Bearing Supports

1 5.4 Structural Details of Bearing Support

15.5 Stopper for Excessive Displacement

1 6. Unseating Prevention Systems

16.1 General

16.2 Seat Length

16.3 Unseating Prevention Structure

1 6.4 Settlement Prevention Structure

1 6.5 Unseating Prevention Structure in Transverse Direction

Appendix : References

1 . References on Earthquake and Active Fault

2. References on Design Earthquake Ground Motion

3. References on Ductility Design Method

4. References on Dynamic Earth Pressure and Dynamic Water

Pressure for Level 2 Earthquake

5. References on Dynamic Analysis

6. References on Liquefaction Evaluation

7. References on Lateral Spreading induced by Liquefaction and

the Design Method

8. References on Strength and Ductility Characteristics of

Reinforced Concrete Columns

9. References on Strength and Ductility Characteristics of Steel Columns

10. References on Strength and Ductility Characteristics of

Prestressed Concrete Superstructures

238
Table 1 Seismic Peformance Matrix

Type of Design Ground Motions Standard Bridges

Oype-A)

Important Bridges

(Type-B)

Level 1 Earthquake: Ground Marions with

Ugh Probability to Occur

SPL 1: Prevent Damage

Level 2 Earthquake:

Ground Motions with

Low Probability to

Occur

Interplate Earthquakes

(Type-D SPL 3: Prevent Critical

Damage

SPL 2 : Limited

Damage for

Function

Recovery

Inland Earthquakes

(Type-D)

Note) SPL: Seismic Performance Level

Table 2 Key Issues of Seismic Performance

SPL Safety Funcuonability

Repairabilty

Short Term Long Term

SPL 1 Prevent Damage

Safty against

Unseating of

Superstructure

Same Function as before EarthquakeNo Need of

Repair for

Function

Recovery

Simple Repair

SPL 2

Limited

Damage for

Function

Recovery

Safety against

Unseating of

Superstructure

Early Function

eroovery can be

made

Function

Recovery can he

made by

Temporary

Repair

Relatively Easy

Permanent

Repair work can

be made

SPL 3

Prevent Critical

Damage

Safety against

Unseating of

Superstructure

239
Overall Goals

Functional Kccjuiremcnts

(Basic Requirements)

Performance

Requirement Level

Verification lvfcthods

and

Acceptable Solutions

Mandated

Specifications

Can be Modifk d or May be Select* sd with

Necessary Vea itications

Fig. 1 Code Structure of JRA Design Specifications, Part V: Seismic Design

Table 3 Applicable Verification Methods of Seismic Performance Depending

on Earthquake Response Characteristics of Bridge Structures

Dynamic

Qiaracteristies

Bridges with

Simple Behavior

Bridges with Xvldti

Plastic Hinges and

withuot Verification

of Applicatn1 ity of

Energy Constant

Rule

Bridges with Limited Application of

Static Analysis

SFL to be

verified

With IS/lulti IVfade

Response

Bridges with

Compl i cated

Behavior

SFL 1

Static

Verification

Static Verification

rjynamic

Verification

rjynamic

Verifiatin

SFL 2/SPL 3 Dynamic

Verification

Ficample of

Bridges

Other Bridges 1)Bridges with

Rubber Bearings

to distribute

Inertia Force of

Superstructures

2) Seasrrically

Isolated Bridges

3) Rigid Frame

Bridges

4) Bridges with

Steel Columns

1) Bridges with

Long Natural

Period

2) Bridge with

Ffigh Piers

1) Gable-stayed

Fridges,

Suspension

Fridges

2) Arch Bridges

3) Curved

Fridges

240
Development of National Guidelines for Seismic Performance Testing

by

W. Phillip Yen1

, Jerry J. Shen2

, and John O'Fallon3

ABSTRACT

In order to make the bridge experiments more

efficient in terms of providing reliable and

comparable information, Federal Highway

Administration has conducted a systematic study

on bridge testing methods. The most common

procedures and issues are identified. Proper

methods on specimen construction, loading

procedure, as well as measurements and data

format will be established to provide

experimental researchers an easy reference that

makes test results comparable to results from

other tests.

KEYWORDS: Bridge performance testing;

earthquakes; guidelines; performance-based

approach; performance testing; Seismic testing

1.0 INTRODUCTION

Experimental studying on bridges and

components of bridges is an essential need for

the improvements on design and construction

techniques. As the seismic design concept for

bridges gradually turns to performance-based

approach, the need for a large amount of

comparable tests results increases. In order to

make the bridge experiments more efficient in

terms of providing reliable and comparable

information, the Federal Highway

Administration (FHWA) has launched a

systematic study on the techniques and

procedures of conducting experimental bridge

ests. The objective of the study is to develop the

National Guidelines for Experimental Tests on

Highway Structures.

A literature review covering a wide range of

bridge tests were performed. Information of the

major experimental research institutes was

collected. Some laboratory administrative

personnel were or will be interviewed to include

their practical experiences. These materials will

be summarized and a proposed experimental

guidelines will be produced. These results will

be made available for experts to review and

comment.

2.0 BENEFITS FROM THE GUIDELINES

Each experimental study has its own purpose

and unique setup to serve this purpose. The

results from experiments always contain more

information than they were designed to provide.

A large part of the information is often either

ignored or not comparable with other tests due

to incompatible format and testing conditions. If

widely adopted experiment guidelines exist, the

part of an experiment that is not essential to its purpose can be conducted in the standard manner. Commonly demanded information will

not be ignored. The result can be used for

theoretical derivation or empirical model

calibration with a solid confidence that the data

were obtained from an experiment compliant to a reliable procedure. In the mean time, the tester receives additional credits and publicities

because the experimental results appear in more

consequent studies.

This guidelines is not similar to some of the

1 Office of Infrastructure, R&D Federal Highway Administration, , 6300 Georgetown Pike, McLean, VA 22101

2 Lendis/FHWA, 6300 Georgetown Pike, McLean, VA 22101

3 Office of Infrastructure, R&D Federal Highway Administration, , 6300 Georgetown Pike, McLean, VA 22101

241
existing testing standard such as those in ASTM

testing manual. Due to the explorative origin of

the tests regulated by this guidelines, making

changes to the procedure in accordance to the

need of the specific test does not immediately

disqualify the results. The intention is to provide

assistance rather than restriction to the

experimental community.

3.0 COMMON METHODS AND PURPOSES

FOR BRIDGE COLUMN TESTS

Dynamic structural tests are commonly divided

into 3 categories. They are quasi-static tests,

pseudodynamic tests, and shake table tests (Kausel, 1998; Dimig et al, 1999). Dynamic

tests on bridges are carried out for either

discovering element properties that were not

clear to engineers or verifying the theoretical

predictions. The most common goals for

dynamic bridge tests are:

(1) To determine strength and ductility

under strong earthquake (capacity)

(2) To determine displacement and

force demand

(3) To determine effectiveness of

retrofit

(4) To observe failure patterns such as cracking, buckling, etc.

(5) To observe the performance of new

design or retrofit methodologies

(6) To develop and verify analytical or empirical models

(7) To inspect existing bridges

Different goal results in different test requirements and test methods. The issues being

addressed in this guidelines are those shared by

a number of different tests.

4.0 GUIDELINES

There are three major issues in a bridge test: specimen, loading, and data handling. The

specimen is a full or reduced replica of the

interested part of a real or imaginary bridge. It needs to be designed to best reflect the bridge

characteristics. The loading procedure is usually a simulation of load history occurs in a real

event. For random-type input such as earthquake

and wind, some load histories that are different

from the real load but possess certain realistic

characteristics are used (e.g. cyclic load). The

results of the tests need to be recorded and

stored with adequate conditioning and format.

4. 1 Model construction

4.1.1 Prototype

For general bridge studies, which may target a group of bridges, the bridge parameters, such as

size, shape, designed loads, and detailing, may

not be available from the original design

agencies. One approach is to conduct a statistical

investigation (Lowes and Moehle, 1995, AboShadi et al, 2000) to determine the most

representative parameters for the group of

bridges. The other approach is to simply use a most common value regardless what group of

bridges the test is designed for. The proper

guideline provisions to these two scenarios are described below:

(1) The National Bridge Inventory (NBI)

Database is a complete collection of

bridge external dimensions, functions,

locations, ages, etc. This information can be converted to that required by the

testers (Lampe and Azizinamini, 2000).

This can largely reduce the time spent

on preliminary investigation and

increase the representativeness of the

selected parameters.

(2) When the tester only needs a general

bridge model, a set of most common

parameters can be provided. For

example, a height of 20 ft (6 m), span of

65 ft (20 m), and width of 40 ft (12 m)

can be provided. The dead load and live

load can be derived from the dimensions

given above. For more convenience

when only one bent is involved, a dead

load of 800 kip (360 ton) and a live load

of 200 kip (90 ton) can be provided. The

sum of dead load and live load

determines the axial load of the column

and the dead load alone determines the

lateral loads.

4.1.2 Scale

242
A dimensionless analysis is used to find the

proper scaling of each parameter (Moncarz,

1981; Bertero, 1984). In the quasi-static bridge

column tests, all time-dependent parameters that

relate to velocity and acceleration are ignored.

The axial load and lateral load from the

superstructure are reduced to an external force

applied vertically and horizontally to the top of

the column. The material properties, such as

elastic modulus, yielding stress, and specific

weight, are kept to the same as the prototype

(scale factor=l). The parameter that needs to be

changed as will is the geometrical dimensions.

The resultant scaling factors for quasi-static tests are listed in the 1

st column of Table I.

Time and time-dependent variables need to be

considered in the scaling for fast tests such as shake table tests. Because of the involvement of

the gravitational acceleration (cannot be changed

in large structural tests), the scaling is diverted

into two courses:

(1) Gravity is insignificant:

The vertical load is either unimportant

to the test or is applied externally. The

self weight of the entire specimen is negligible to the purpose of the test. Supplementary vertical load (if

necessary) is applied through force

actuators. The input and response

acceleration is allowed to be scaled

without changing specific weight. The

proper scaling is shown in the 2

nd

column of Table I.

(2) Gravity is essential:

If the gravity force from the self weight

is one of the important factor in the test,

the horizontal acceleration needs to be

kept the same scale as the gravitational

acceleration (unity) because the vertical

(gravity) force and horizontal (inertia)

force come from the same mass. Due to

the difficulty of changing specific

weight in specified proportion, the

scaling of specific weight is substitute

by auxiliary masses. The auxiliary mass

needs to be attached to the heaviest

places, which is usually the

superstructures, without introducing

additional stiffness. The proper scaling

and auxiliary mass are listed in the 3

rd

column of Table I.

It has been noticed that the construction

materials have different mechanical properties at different size. The mechanical properties such as

elastic modulus and yielding or ultimate stress are not only difficult to change by any specific

scale but also difficult to maintain constant when

size is changed. There are multiple reasons that

make the mechanical properties size-dependent,

such as different manufacturing process and

failure mechanism. Special care should be taken

to ensure this does not alter the test result

unexpectedly.

4.2 Load histories

The cyclic loading procedure is usually

considered to be a good simulation tool for

earthquakes when more realistic options, i.e. true

dynamic tests, are not feasible. Since it is an

approximation of earthquake load, there can be

numerous approaches. Most tests done to bridge

columns have adopted a progressive pattern

cyclic loading procedure (Figure 1). This

procedure involves some pilot cycles in the

elastic range that provides more accurate

estimation for yielding displacement, followed

by plastic cycles that contain 2-3 cycles at each

deformation level and increase in multiples of

yielding displacement. It is believed that two or

three cycles for each displacement level is a balance between demonstration of strength

degradation and avoidance of undesired early

fatigue fracture (Lowes and Moehle, 1995). This

procedure was evolved from the standard steel

element cyclic procedures (ECCS, 1986; Stone

and Cheok, 1989; ATC-24, 1992). This

procedure provides a general test condition for

earthquake loads.

When specific earthquake record is to be tested

on the specimen, a preliminary analysis is required to estimate structural response, and

consequently the displacement history of the

specimen (El-Bahy et al, 1999).

The shake table test is a real dynamic test. As

being more realistic, it is important to be careful

controlling the test environment. The loading

records need to be properly filtered and scaled.

243
Elastic tests should be conducted prior to full- scale nonlinear tests. Modal parameters are

retrieved in the elastic tests in order to well

categorize the model. One or more of the

hammer test, harmonic input test, and random

vibration test can be used for this purpose.

Pseudodynamic tests represent the balance of

actuality and test facility limit. In the slow

pseudodynamic tests, the inertia force and

viscous damping force are calculated in a computer that controls the loading procedure.

The design of the real part and imaginary part as well as the scaling follows the specimen

construction part of this guidelines.

4.3 Measurements and data acquisition

The technologies used in measurement systems

evolve quickly. New issues emerge while old

issues are resolved. It is not easy to setup

general guidelines for all tests. However, wellconfigured measurement and acquisition

systems share some similar principles. It is important to layout the principles to be followed.

A few fundamental values, such as lateral load,

lateral displacement, and curvature, should be

required by the guidelines. Some physical detail

that may affect the resultant accuracy should be

specified. For example, the range of the

curvature measurement (with respect to section

size) needs to be sufficient to cover the plastic

hinge zone and part of the elastic zone. Any

potential slippage surface needs to be monitored on both sides.

The data acquisition systems have all become

digital nowadays. Analog filtering is crucial for

such systems. For slow static tests, the system

can stop at each designated load step and allow

the measurements recorded. Analog filters can be conditionally absent if several readings are made and averaged to eliminate the fluctuation

from high-pitch electronic or physical noise. For

high-speed tests, the analog low-pass filtering

range needs to cover the highest frequency of

the interested vibration component. The

sampling rate should be higher than twice of the

filter cut-off frequency.

The study on data format, storage, and

transmission is currently one of the major efforts

from many experimental institutes lead by the

National Science Foundation (NSF). The

progress will be closely observed and integrated

with this guidelines.

5.0 SUMMARY

Most bridge column tests are either for research

purpose or for specific construction or retrofit

project. The uniqueness of these tests makes it

difficult to set up general guidelines for all. However, as a leading agency in the highway

industry, the FHWA will persevere in the effort

of laying out the national testing guidelines for

bridges due to the anticipated tremendous

benefits to the researchers and designers in this

country. The first appearance of the product is expected to take place in a short time. This

project will not be able to provide premium

result without the contributions from bridge

testing experts. All suggestions and discussions

are highly appreciated.

6.0 REFERENCES

[1] Abo-Shadi, N.; Saiidi, M.; Sanders, D. July

2000. "Seismic Response of Reinforced

Concrete Bridge Pier Walls in the Weak

Direction," Technical Report MCEER-00-

0006.

[2] Applied Technology Council. 1992.

"Guidelines for Cyclic Seismic Testing of

Components of Steel Structures," ATC-24,

Applied Technology Council, Redwood

City, CA.

[3] Bertero, V. V.; Aktan, A. E.; Charney, F. A.; Sause, R. 1984. "Earthquake Simulation

Tests and Associated Studies of a l/5thscale Model of a 7-story R/C Frame-wall

Test Structures," Report No. UCB/EERC84/05, University of California, Berkeley.

[4] Dimig, J.; Shield, C, French, C; Bailey, F.; Clark, A. Sept. 1999. "Effective Force

Testing: A Method of Seismic Simulation

for Structural Testing," Journal of

Structural Engineering. ASCE, 125(9), pp

1028-1037.

244
[5] El-Bahy, A.; Kunnath, S.; Stone, W.;

Taylor, A. September 1999. "Cumulative

Seismic Damage of Circular Bridge

Columns: Variable Amplitude Tests," ACI

Structural Journal, 96(5), pp 71 1-719.

[6] European Convention for Constructional

Steelwork, Technical Committee 1,

Structural Safety and Loadings; Working

Group 1.3, Seismic Design. 1986.

"Recommended Testing Procedure for

Assessing the Behavior of Structural Steel

Elements under Cyclic Loads," First

Edition.

[7] Kausel, E. May 1998. "New Seismic

Testing Method. I: Fundamental Concepts,"

Journal ofEngineering Mechanics, ASCE,

124(5), pp 565-570.

[8] Lampe, N.; Azizinamini, A. Nov. 2000.

"Steel Bridge System, Simple for Dead

Load and Continuous for Live Load," Proc.

Conference of High Performance Steel

Bridge, Nov. 30~Dec. 1 2000, Baltimore,

Maryland.

[9] Lowes, L., N; Moehle, J., P. Sept. 1995.

"Seismic Behavior and Retrofit of Older

Reinforced Concrete Bridge T-Joints,"

Report No. UCB/EERC-95/09, Earthquake

Engineering Research Center.

[10] Moncarz, P. D. 1981. "Theory and

Application of Experimental Model

Analysis in Earthquake Engineering,"

Dissertation, Stanford University.

[11] Stone, W.; Cheok, G. Jan. 1989. "Inelastic

Behavior of Full-Scale Bridge Columns

Subjected to Cyclic Loading," NIST

Building Science Series 166.
TABLE I SCALING FACTORS FOR DIFFERENT TESTS

Variable Slow test Dynamic test

Gravity insignificant Gravity essential

Geometric size { SL sL sL Time t N/A s,

OLICdO (y 1i 1 1 Strain e 1 1 1 Elastic modulus E 1 1 1

Yielding stress CTy 1

1 1 Density p N/A 1 1 Force P Su2 Su

2 sL2

Bending moment M sL3 SL3 sL3

Rotation angle (a/L) q 1 1 1 Curvature § SL

"

1 SL

" 1 SL"

1

Displacement U SL SL SL Acceleration N/A SL

" 1 1

Auxiliary mass N/A 0 SL2-SL

3

External damping N/A sL

2 sL1 - 5

External stiffness sL SL

Figure 1 The progressive cyclic load history

246
SPECIAL THEME 1

TERRORISM: COLLAPSE of the

WTC and PENTAGON WEDGE A

Terrorist Threats Against the Pentagon

by

Robert L. Hall

1

, James T. Baylot

1

, Robert J. Dinan1, Richard Dove1

, John Hayes2, W. Matthew Hossley1

, James L. O'Daniel1

, Thomas R. Slawson1

, Stanley C. Woodson1

ABSTRACT

The terrorist attack of 11 September 2001

against the Pentagon prompted USACE to perform a 30-day study to assist in the

development of anti-terrorist methods for the

Pentagon structure. Drawing on years of

experience and model development, USACE

provided and analyzed structural concepts for

the rebuild and renovation of the Pentagon.

USACE personnel assisted in the damage

assessment immediately following the terrorist

event, analyzed the current configuration and

design concepts of the Pentagon wall system

using simplified engineering-level and high

performance computational models, and

developed strength measures that were inserted

into a PC-based vulnerability assessment code.

Recommendations were provided to the

Pentagon Renovation Program.

years, experimental and analytical studies have

resulted in significant improvements in the

understanding of the response of conventional

aboveground structures to blast effects. The

studies have led to the development of

guidelines and computer software for assisting

engineers in evaluating and designing protective measures.

At the time of the 11 September attack, the

Pentagon was undergoing a renovation that

included retrofit measures to increase the

resistance of the structural envelope to terrorist

attack, particularly for blast effects. Although

the Pentagon Renovation Project took great

advantage of state-of-the-art technology

available at the time, this technology is steadily

advancing. The opportunity now exists to incorporate recent advances into the areas to be

reconstructed, as well as areas not yet renovated.

KEYWORDS: blast effects; Pentagon

renovation; retrofit; structural assessment;

terrorist attack; vulnerability

1.0 INTRODUCTION

Following the 11 September 2001 terrorist

attack on the Pentagon, USACE entered into a 30-day study agreement with the Pentagon

Renovation Program managers to assist in the

evaluation and development of physical antiterrorist measures for the Pentagon. USACE is uniquely qualified to blend state-of-the-art

research, simple and sophisticated analytical

techniques, and practical engineering into a

focused study to provide concepts for antiterrorist measures. The bombing of the Marine

Corps barracks in Lebanon in 1984 was the

motivation for the U.S. Army to establish one of

the first research programs to address the effects

of terrorist weapons on conventional buildings

and develop improved design and retrofit

methods for these facilities. During the past 15

2.0 PURPOSE

The purpose of the 30-day study was to examine

protective measures for the Pentagon for a range

of potential types of threats and threat levels that

include airblast from explosive detonations, fire

hazards, and chemical/biological/radiological

weapons. The focus was general protection for

all building occupants, rather than localized

protection for specific critical assets. The 30-

day study was not an effort to design for specific

performance in response to a specific defined

threat. This paper will summarize the blast

response of the unretrofited. Pentagon and the

added protect provided by the existing retrofit

used in wedge 1

.

3.0 11 SEPTEMBER 2001 EVENT

U.S. Army Corps of Engineers personnel

conducted a Pentagon damage assessment from

14-17 September 2001. The overall purpose of

this initial forensic study was to investigate the

249
extent of the damage caused by the plane that

crashed into the Pentagon on 1 1 September. An

important part of this effort was gathering

information on the performance of the retrofitted

components of the Pentagon. While, the recent

retrofit upgrade techniques employed in the

building's renovated sections were intended to increase protection against vehicle bombs, the

response of the retrofit components under the 1

1

September loads provides insight into the overall

effectiveness of the upgrades. The structure

experienced the extraordinarily energetic

localized impact of the plane itself, secondary

impacts, and the complex external/internal blast

and fire loads from the fuel carried within the

plane. While the complexity of the loading

complicates a quantitative determination of the

exact load/response relationship, a qualitative

assessment of the relative response of the retrofit

components as compared to the non-retrofit

components is feasible. This is possible because

the plane struck close to the interface between

the retrofit and the non-retrofit building sections.

Figure 1 is a schematic of the layout of the

Pentagon. Wedge 1 consists of sections 3 and 4

as shown in this figure. The impact location and

orientation are also indicated on the figure. The

exterior wall near the impact location is shown

as Figure 2. The boundary of Wedge 1

(retrofitted) with Wedge 2 (non-retrofitted) is shown near the left side of the figure. The

impact occurred near column line 14 of Wedge

1. The collapsed region (Figure 3) extended

from the expansion joint near column line 1 1 to column line 18. Windows near the Wedge 1

- Wedge 2 boundary are shown in Figure 4.

Damage to the non-retrofitted windows is higher

than damage to the retrofitted windows,

although the retrofitted windows are closer to the impact area and subsequent fuel fire/blast.

The retrofit of Wedge 1 was effective in saving

lives during the events of 11 September 2001.

While, the retrofit upgrades were intended to increase protection against vehicle bombs, they

appeared to have been effective in mitigating

some of the secondary effects of the plane

impact. The retrofit windows responded in a

ductile and tough manner to the blast loads and

were also able to resist substantial fragment

impacts. The windows also remained intact even when apparently engulfed in the fireball of

burning fuel, preventing entry of the fire into the

upper floors. The tube steel members worked as intended and may also have played a part in the

resilience of the building, a possibility that was

not evaluated in this study. The annotated

photographic documentation of the response of

both the non-retrofit and retrofit windows may

prove useful in future analytical and

experimental efforts to improve protection

against terrorist attack.

4.0 UN-RETROFITTED PENTAGON

CONFIGURATION

The unretrofitted wall analyzed consists of 5-1/2

inches of limestone facade with an 8-in., two

wythe, unreinforced brick infill wall. The floor

height varies but is taken as a nominal 10 ft for

this analysis. The horizontal span between

columns is 10 ft on-center. The construction

quality of the edge support conditions of the

masonry wall (gap between wall and columns or grout fill) is uncertain, but it is assumed that the

wall is supported on all 4 sides (two-way action)

without the support rigidity that could induce

arching (compressive membrane action). The

limestone facade was assumed to respond in

one-way action between column lines. In

addition, nearly every bay except those on the

fifth floor has a window. The window analyzed

is 7-feet high by 5 -feet wide (outside

dimensions), with Vi-inch thick annealed glass.

5.0 EXISTING (WEDGE 1) RETROFIT

The retrofits currently used in the Pentagon

renovation (Figure 5) include blast hardened

windows, supported by horizontal tubes that

frame into vertical tubes that run from floor slab

to floor slab. The 6"x6" or 8"x8" tubes transfer

the dynamic reactions from the windows and the

exterior masonry walls to the floor slabs. The ERing retrofit window for the Wl type consists of an insulated, or double pane glazing. The outer

pane is lA in. thick thermally tempered glass,

with a Vi in. air gap. The inner pane is a laminate consisting of 3/8 in. thermally

tempered glass, .090 in. PVB inner layer, and

3/8 in. thermally tempered glass. In addition, a

250
geotextile membrane is used over the interior

surface of the masonry wall to prevent the

masonry from becoming a debris hazard during

a blast event.

6.0 ANALYSIS

The analyses assume that the Pentagon frame is

sufficient to resist the loads transferred to it from

the exterior walls. Evaluations of the original

and retrofitted Pentagon structures were

performed using the Antiterrorist (AT) Planner

software [1]. AT Planner is a PC-based

computer code that assists installation-level

personnel in analyzing the vulnerability of

buildings and their occupants to the effects of

terrorist vehicle bombs. The program also

contains information to aid in developing

protective measures.

AT Planner is being developed to present

concepts and procedures for protecting

deploying forces from terrorist/saboteur attack

using expedient methods that require a minimum

of engineer resources. Recent experience has

shown that the demand for military engineering

in support of antiterrorism has risen dramatically

as the Army is drawn into a succession of

operations other than war. In these situations,

U.S. troops may be subject to attack by

unfriendly civilian or paramilitary groups. AT

Planner is a Windows 95 -based application

suitable for operation on a notebook computer

by combat engineer officers, and draws on

completed and ongoing research related to the

protection of fixed facilities from terrorist attack

as well as work on field fortifications. AT

Planner is based on references 2-7. AT Planner

provides standoff distance evaluations, structural

damage and window hazard calculations,

protective measures checklist for terrorist

threats, and vehicle velocity calculations and

barrier recommendations. When a vulnerability

analysis from a terrorist bomb is calculated in

AT Planner, blast pressure is calculated at the

center of each structural bay on a structure. Angle of incidence is considered in calculating

airblast levels on structures, but clearing effects

and shielding effects are not. AT Planner uses PI (Pressure Impulse) diagrams to allow a user

to quickly estimate building damage from a

vehicle bomb attack. A description of a PI curve

is shown in Figure 6. The hazard/protective

levels used by AT Planner in this study are shown in Figure 7. AT Planner has a set of

default damage diagrams and the capability to read in user specified diagrams. The AT Planner

default building components (excluding

windows and non-reinforced masonry walls) are normalized so that the user can modify the span

and thickness of the component. The user

specified diagram is not normalized; therefore a

different diagram must be developed for each

different building layout. For this reason, a

significant amount of time was spent studying

construction diagrams in order to determine

nominal specifications before the damage

diagrams were developed.

Wall response calculations were performed

using the Wall Analysis Code [8] (WAC).

WAC is a single-degree-of-freedom (SDOF)

model designed specifically to calculate the

response of reinforced and unreinforced concrete

and masonry walls to airblast loads. Window

response/hazard predictions were made with

HAZL [9]. These response predictions were

used to develop the PI diagrams that were input

into AT Planner to perform the blast evaluations.

The PI diagrams for the walls were validated by

performing analyses using higher-fidelity, finite

element (FE) structural response models

described in Appendix 3.D.

Range-to-effect (RTE) and PI curves for the

Ring E walls and windows were developed and

entered into AT-Planner. In addition, PI curves

for the reinforced concrete roof panels required

to perform AT Planner assessments. Roof panel

retrofits were not required for blast protection in

subsequent scenarios - the curves used for the

roof panels are constant throughout this report.

RTE curves for this report are shown for

components exposed to normally reflected loads.

They are not shown for roof panels since

realistic roof loads are closer to incident

pressures rather than normally reflected

pressures. The damage levels obtained from the

supporting FE models match well with the PI

diagrams developed using WAC, thus validating

the PI diagrams. The PI curves presented above

are used in AT Planner to define safe stand-offs

251
around the Pentagon for the large and small

truck bomb threats as shown in Figure 8 (the

windows control these stand-offs).

To analyze the existing retrofits response to blast

load, SDOF models of the wall and window

systems were developed. The wall model did

not consider the effects of window failure. The

resistance of the wall included the strength of

the facade, the masonry wall, and the tubular

framing system (dominant contribution). The

wall system model was used in WAC to

generate RTE and PI curves and these curves were validated with FE analyses. The high level

PI curves were used in AT Planner to define safe

stand-offs around the Pentagon for the large and

the small truck bomb threats. The custom PI

diagrams for the window and wall retrofits of

the exterior wall of the E-Ring were used for all walls. Damage plot in figure 9 are intended to

illustrate damage to the outside of the E-Ring

only.

7.0 RECOMMENDED RESEARCH/

EVALUATION

Doors in the exterior wall of the E-Ring need to be designed to resist the same bomb size -

standoff combination that the E-Ring exterior

walls and windows are designed to. Other

components of the exterior shell of the Pentagon

need to be evaluated for the same conditions

used to design the E-Ring exterior. Components

that need to be evaluated include the walls

windows and doors on the back of the E-Ring

and on the backs and fronts of the A-, B-, C-,

and D-Rings. The roofs of the two-story

portions of the buildings located between the Dand E-Rings and between the C- and D-Rings

also need to be evaluated.

Further blast hardening should be considered in

conjunction with proposed plans to relocate

public roads and/or further control access to the

Pentagon roads and parking areas. The

Pentagon should be evaluated against other

reasonable threats. The wall/window retrofit or

rebuild selected for use in the Pentagon

rebuild/restoration should be tested to verify the

level of protection provided.

8.0 REFERENCES

1] AT Planner, U.S. Army ERDC, Vicksburg,

MS, October 2001.

2] FM 5-114, Engineer Operations Short of

War, Headquarters, Department of the Army,

13 July 1992.

3] TM5-853-2/AFMAN 32-1071, Security

Engineering Manual, Vols. 1-4, Departments of

the Army, and the Air Force, May 1994.

4] "Structures to Resist the Effects of

Accidental Explosions," ADA 243 272/TM

5-1300/NAVFAC P-397/AFR 88-22,

Departments of the Army, the Navy, and the Air

Force, November 1990.

5] Drake, J.L., et. al., "Protective Construction

Design Manual," prepared for Air Force

Engineering & Services Laboratory, Tyndall Air

Force Base, Florida, November, 1989, ESL-TR87-57.

6] Ferritto, J., Tancreto, J., "User's Guide for

Navy Air Attack Simulation Program

(NAASP)," Naval Facilities Engineering

Command, Alexandria, VA, Program No.

Y0995-00 1-06-0 10, May 1987.

7] Oswald, C.J., and Skerhut, D., "FACEDAP

Theory Manual," Contract No. DACA45-91-D0019, U.S. Army Corps of Engineers, Omaha

District, April 1993.

8] Slawson, T.R., "Wall Response to Airblast

Loads: The Wall Analysis Code (WAC),"

prepared for the U.S. Army ERDC, Vicksburg,

MS, Contract DACA39-95-C-0009, ARA-TR95-5208, November 1995.

9] "Window Fragment Hazard Level Analysis,

the HAZL Model," U.S. Army ERDC,

Vicksburg, MS, October 2001.

9.0 ACKNOWLEDGMENT

The authors greatly appreciate the efforts of all members of the study team and the permission

to publish granted by the Director, Geotechnical

and Structures Laboratory.

252
Figure 1 . Layout of Pentagon showing impact location.

Figure 2. Exterior of E-Ring of Pentagon Wedge 1 and 2 with column line designations. Wl designation

on windows denotes non-operable (fixed) retrofit window. Impact occurred near Column Line 14,

collapsed region extended from construction joint at Column Line 1 1 to Column Line 18.

Geotechnical and Structures Laboratory, Engineer Research and Development Center,

U.S. Army Corps of Engineers, Vicksburg, MS 39180

Construction and Engineering Research Laboratory, Engineer Research and Development

Center, U.S. Army Corps of Engineers, Champaign, IL 61826

253
Figure 3. Collapsed section/plane impact area in Wedge 1. Plane entered the 1

st floor at an angle of

approximately 40 degrees from perpendicular moving toward the left into Wedge 2.

Figure 4. Exterior view of transition between Wedge 1 (on right, retrofit) and Wedge 2 (on left, nonretrofit). First floor non-retrofit window adjacent to transtion may be High Hazard blast response.

Second and 3

rd floor fire damage. Approximately 135 ft (44.2 m) North of plane impact point.

254
Figure 5. Current retrofits.

Pressure Impulse Curve

A pressure-impulse (PI)

diagram is an iso-response curve

for a particular structure loaded with

a particular load history type.

1000

100

0)u>3

a

E

10

1.0

The PI curve divides the plot into

two regions: (1 ) the area above and

to the right where the response level impijise Asymptote

of the PI curve is exceeded and

(2) the area below and to the left. 01

Pressure Asymptote Pressure

At low impulse (short duration loads), the response is essentially

independent of peak pressure and depends only on the impulse.

Quasi-Static Loading

Dynamic Loading

^/Impulsive Loading

0.1 1.0 10 100 1000

At low pressures (long duration loads), the response is independent

of impulse and depends only on the peak quasi-static load.

Figure 6. PI curve description.

255
1000

100 — Lower Limit of Low Damage — Lower Limit of Medium Damage - Lower Limit of High Damage

1000

No damage: No appreciable damage; the component is reusable without repair. This damage level can be equated with a High Level of Protection.

Low damage: The component is probably repairable and it has provided a generally adequate level

of protection to personnel and equipment from the effects of the explosion. This damage level can

be equated with a Medium Level of Protection.

Medium/Moderate Damage: Repair of the component is not feasible, but it has not collapsed, and it has provided substantial protection to personnel and equipment from the effects of the explosion.

This damage level corresponds to the greatest degree of damage that might be accepted, thus it can

be equated with a Low Level of Protection.

High Damage: The component is definitely beyond repair but it has not necessarily completely

collapsed. It has undergone a deformation such that it cannot be counted on with high certainty to protect personnel and equipment from the effects of the explosion. This damage level is equated

with "Collapse" as it is used in terms of a Level of Protection. However, components with 100%

damage will most probably not be collapsed in the general usage of this word.

Figure 7. Hazard/Protection Levels.

256
Small Bomb

Large Bomb

Figure 8. Stand-offs to prevent high hazard for the large and small truck bombs for the unretrofitted

Pentagon.

Threat location on

High Hazard Distance

Figure 9. Stand-offs to prevent high hazard for the large and small truck bombs for the

current Pentagon retrofits.

257

SPECIAL THEME 2

DISASTER INFORMATION

Natural-Technologic Events: The Frequency and Severity of Toxic Releases During and After

Natural Disasters with Emphasis on Wind and Seismic Events

by

Stacy Young1

ABSTRACT

Toxic substances released from storage

facilities, industrial plants, or transportation

vehicles as the result of technologic emergencies

(e.g., equipment failures, severed pipelines)

precipitated by natural disasters are known as natural-technologic or na-tech events. The risks

associated with na-tech discharges are of

importance given accelerated technologic and

industrial expansion, increasing population

density in disaster-prone areas and trends in the

more frequent occurrence of natural disasters.

These trends increase the probability of

catastrophic future disasters and the potential for

mass human exposure to toxic agents released

during disasters. Half of the natural disasters

that occurred in the United States and its

territories during the 1990s resulted in at least

one na-tech event. We identified and

characterized 1,152 na-tech events that occurred

during the 1990s. Of these, 580 were related to

wind or seismic effects. Greater knowledge of

the factors that resolve the occurrence and

severity of na-tech events can be used to increase disaster preparedness levels and

damage adsorption capacities.

KEYWORDS: Chemicals; Exposure; Health

Effects; Na-tech; Natural Disasters; Toxic

1.0 INTRODUCTION

Subterranean stress, surface instability, high

winds, and abnormal precipitation or temperature are natural circumstances that

threaten human health (/). Rapidly moving or

rising floodwaters, damaged roadways, and

collapsing buildings are hazards commonly

associated with extreme natural events. Oil

1 National Center for Environmental Health, Centers

for Disease Control and Prevention, 1600 Clifton

Road, NE (Mailstop E-23), Atlanta, GA 30333 USA

releases, agrochemical pollution, asbestos dust

and other toxic materials released during

extreme natural events can also endanger human

health. Toxic releases that result from

technologic emergencies created by natural

disasters have been referred to as naturaltechnologic or na-tech events (10).

Na-tech releases pose potential threats to human

health through both acute and chronic exposures

that may occur when environmental integrity is compromised. Oil, chemical, radiological, or

biological agent releases may pose extreme fire

and explosive hazards, result in the formation of

toxic vapors, or lead to contamination of

waterways and groundwater sources (8, 11, 12).

After a 1995 quake in Kobe City, Japan, dust

and irritants raised during demolition work were

considered a factor in the deteriorating condition

of asthma patients (7). In the 1989 Loma Prieta

earthquake near Santa Cruz, California,

hazardous materials exposures accounted for

20% of after-earthquake work-related injuries

(3), and numerous human exposures to toxic

materials (e.g., asbestos and fiberglass

insulation, mercury, leaking transformers,

broken chemical containers) (6).

A lack of standardized record keeping, however,

has hampered efforts to assess the frequency and

severity of these occurrences (10). As a result,

the significance of hazardous material releases

in the broad scope of disaster-associated risks

has yet to be determined. Na-tech events are a preventable consequence of natural disasters.

Na-tech occurrence is the result of technologic

circumstances in combination with a natural

disaster. Though little can be done to modify

the nature of a disaster, the technologic

component offers a point of intervention that can be used to diminish or eliminate toxic releases

resulting from na-tech causes in disaster

situations. In this study, we present estimates of

the frequency and character of na-tech incidents

261
that occurred in the United States and its

territories during the 1990s—with emphasis on

wind- and seismic-related na-tech events.

Characterization of these events will aid disaster

preparedness and mitigation.

2.0 METHODS

We used National Response Center (NRC)

records to assess oil, chemical, radiological, and

biological discharges occurring in the United

States and its territories from January 1, 1990,

through December 31, 1999. We used Federal

Emergency Management Agency (FEMA)

records to identify climatological and

geophysical events that occurred in the United

States during the same period and resulted in

Emergency Declarations or Major Disaster

Declarations2. A toxic release was identified as

a na-tech release if all of the following criteria

were met: (1) cause of the release was a "Natural Phenomenon," (2) release day fell within a disaster period, (3) state of release and

state of FEMA declaration matched, and (4)

county of release and county of FEMA

declaration matched. Toxic releases attributed

to non-natural disasters (e.g., the Oklahoma City

bombing) and fire events were removed from

analysis.

We characterized na-tech events with respect to: type of natural incident, type of declaration, year

of occurrence, type of facility/mode of

transportation, location of release (e.g., EPA

region), total amount of discharge, and chemical

class.

3.0 RESULTS

The NRC recorded 275,032 toxic incidents

during the 10-year study period. Of these,

177,464 had sufficient information on the cause,

An Emergency Declaration can be declared when

the President determines Federal assistance is needed

to supplement State and local efforts in providing

emergency services. A Major Disaster Declaration

may be presidentially determined for any natural

catastrophe resulting in severe damage exceeding the

combined capabilities of State and local governments

to respond.

date, and location of the toxic release to

facilitate identification of na-tech events.

FEMA records indicated that 480 natural, nonfire disasters occurred in the United States and

its territories during the same period. Over half

of the natural events (52.0%) were associated

with at least one na-tech release. The remaining

events (47.3%) resulted in no na-tech releases.

The number of na-tech releases per event ranged

from 0 to 42 with an average of 2.4 na-tech

releases per natural event. A total of 1,152 na- tech events were identified.

3.1 General Na-tech Characterization

Nearly half (48.2%) of all na-tech events were

generated by wind-related storms, with floodrelated storms (30.5%), severe winter storms

(19.2%) and earthquakes and aftershocks (2.2%)

accounting for the remainder (Figure 1). Further

analysis presented in this report focus solely on

the 580 na-tech events attributed to wind and

seismic events during the 10-year period, unless

otherwise noted.

The majority of wind- and seismic-related na- tech events were associated with a major disaster

declaration (98.5%). Na-tech events were

reported more frequently during the latter half of

the decade (61.2%) than during the first half

(38.8%). Nearly 73% of all discharges were

from fixed facilities (e.g., factories, laboratories)

with marine-related and pipeline-associated

discharges accounting for 14.5% and 8.8% of

releases, respectively. EPA Regions TV and VI

experienced the greatest number of na-tech

releases—190 and 175, respectively (Figure 2).

3.2 Annual and Geographic Na-tech Fractions

and Ratios

The crude na-tech fraction for the period from

1990 through 1999 was 2.1 per 1,000 releases

(95% confidence interval [CI]: 2.0-2.2 per 1,000

releases). Analysis of na-tech fractions in 2-year

intervals suggested that na-tech occurrence

increased during the last two years of the decade

(Figure 3). Geographically, the highest na-tech

fractions of release occurred in EPA regions rV

(4.60 [95% CI: 4.57-4.63]), I (2.37 [95% CI:

262
2.36-2.38]), DC (1.97 [95% CI: 1.97-1.98]), and

VI (1.87 [95% CI: 1.87-1.88]).

Regions VI (7.3 [95% CI: 5.4-9.9]) and DC (2.9

[95% CI: 2.5-3.5]) exhibited the greatest number

of na-tech events per natural event. Earthquakes

were associated with 5.00 [95% CI: 4.0-6.3] natech events per natural events and wind-related

storms were associated with 2.7 [95% CI: 2.4-

3.1] na-tech events per natural event.

3.3 Amounts of Toxics Released

In 580 na-tech events, 651 toxic materials were

released. Measures of the total volume released

were available for a limited number of the 580

na-tech events. Earthquakes and aftershocks

(n=10) produced a greater median release

volume (3880.0 L) than wind-related storms

(n=288; 113.6 L). Release volumes and masses

during the period from 1990 through 1994

(median=310.4 L; 453.6 kg) were greater than

those between 1995 and 1999 (median=111.7 L;

102.1 kg).

3.4 Types of Toxics Released

Of the 651 hazardous materials released during

wind- or seismic-related na-tech events, 465

could be characterized on chemical class. The

majority of discharged chemicals (324 [69.7%]),

were flammable liquids.

4.0 DISCUSSION

Previously, anecdotal accounts and regional

surveys lent evidence to the existence of na-tech

events during and after natural disasters (2, 9,

13, 14). Our identification of 580 wind- and

seismic-related na-tech events during the 1990s

reinforces anecdotal descriptions and regionally

conducted surveys in demonstrating the

existence of na-tech threats.

A 1994 survey of hazards mitigation

organizations in 20 states showed an increase in

the number of na-tech events during the 1980s

(10). Our data suggest that wind and seismic natech occurrence increased during the 1990s.

However, our ability to definitively note trends

and make comparisons between groups is limited by the quality of the data.

The NRC data are based on initial release

accounts made during or immediately after an

incident when exact details may be unknown.

As a result, only information that is available

early on in the incident is available. Rarely do

these early reports indicate the exact amounts

released or whether injuries or deaths resulted

from a release. Inaccurate, insufficient, or missing NRC data limited the quality of some

data and continued to be a problem during our

na-tech identification process. One-third of all discharges reported to the NRC did not have

sufficient information to allow for determining

whether they were na-tech discharges. Voids in

the data limit our ability to identify trends and

make comparisons, but may also increase the

likelihood that the approximations of na-tech occurrence presented here are underestimates of

the true occurrence.

Despite the limitations of our study, we

identified 580 wind and seismic na-tech events

that occurred during the 1990s. In our data,

natural events of all types had the potential to

result in a na-tech disaster. Windstorms

produced the greatest number of na-tech events

and earthquakes resulted in the greatest

quantities of release. Wind-related storms and

earthquakes also had the largest ratios of na-tech

events per natural event—indicating that how

often some types of natural events occur is not

the sole factor determining their relevance with

respect to na-tech events. Earthquakes have

previously been found to pose the greatest na- tech threat (10).

While the frequency of na-tech occurrence appeared to increase during the 10-year period,

the median volume and mass of releases

decreased. One possible explanation for this

paradoxical finding may be that safety measures

put in place during the 1990s successfully

limited large releases and heightened awareness of na-tech threats, thus increasing the reporting

of smaller releases.

5.0 CONCLUSION

263
Our study reported the occurrence of wind- and

seismic-related na-tech events during the 1990s,

but was unable to identify significant trends in occurrence. The frequency of na-tech events

during the study period suggests that na-tech

events may be more pervasive than commonly

thought and could be occurring with increasing

frequency (4, 10).

Prevention of future na-tech events is the

ultimate goal. Some preventative actions that can be taken are identifying and addressing

known hazards, providing adequate information

and skills to individuals most likely to be

exposed or in-contact with toxic agents during or

after natural disasters, and increasing municipal

and private awareness of the environmental

hazards associated with natural disasters. Our

study aids such prevention efforts through

characterizing the circumstances by which

historical na-tech events have occurred in an

effort to identify future hazards. The study is also an effective tool for establishing that toxic

exposures are occurring in the wake of natural

disasters—potentially increasing awareness of

disaster-associated environmental hazards and

spurring the development of appropriate

mitigation and preparedness plans in disasterprone states and countries that might otherwise

have difficulty generating interest in such

activities.

Modifying the NRC database or developing a new surveillance tool would yield more useful

information on na-tech frequency and more

detailed accounts of incidents. Thoroughly

characterizing na-tech events will aid in

providing communities with accurate and

appropriate information so that preparedness

levels and damage adsorption capacities can be

raised in an effort to thwart future impacts (5).

5.0 REFERENCES

1 . S. Binder, L. M. Sanderson, Annals of

Emergency Medicine 16, 1081 (1987).

2. K. Breslin, Environmental Health

Perspectives 101, 590 (1993).

3. M. E. Durkin , C. C. Thiel , J. E.

Schneider, T. DeVriend, Bulletin ofthe

Seismological Society ofAmerica 81,

2143(1991).

4. T. Fields, Natural Hazards Observer 24,

1 (2000).

5. Guha-Sapir, M. F. Lechat, Disasters 10,

232 (1986).

6. R. Nathan, K. R. Olson, G. W. Everson,

T. E. Kearney, P. D. Blanc, Western

Journal ofMedicine 156, 278 (1992).

7. J. Nukushina, Epidemiologia e Prevenzione 19, 226 (1995).

8. Proceeding of the Hazmacon

Conference, Santa Clara, CA, April 18-

20, (Dames & Moore, Los Angeles, CA,

1989).

9. L. M. Sanderson, in Hazardous

Materials Toxicology: Clinical

Principles ofEnvironmental Health, J. B. Sullivan, G. R. Krieger, Eds.

(Williams, Baltimore, Maryland,

1992), pp. 326-331.

10. P. S. Showalter, M. F. Myers, Risk

Analysis 14, 169 (1994).

11. K. V. Steinbrugge, H. J. Lagorio, J. R.

Davis, J. H. Bennett, G. Borchardt,

Geology 397, 153 (1986).

12. J. W. Stratton, in The Public Health

Consequences ofDisasters, M. B.

Gregg, Ed. (United States Department

of Health and Human Services,

Atlanta, 1989), chap. 4.

13. K. J. Tierney, Urban Resources 5, 33

(1989).

14. K. White, Environmental Health

Perspectives 101, 584 (1993).

264
Figure 1. Type of natural incident associated with na-tech events
Figure 2. Two-year total NRC toxic releases and na-tech fraction of release per 1,000 NRC releases

(95% CI)

70000 T

« 60000

o>w

£ 50000

06

.g 40000

o

y 30000

gZ 20000

o

H 10000 -

+ + +

7

T 6

5 §

4 4>a

o

3 50

2 ^

u

I

1 « 1 z

1990-1991 1992-1993 1994-1995 1996-1997 1998-1999

266
Disaster Information Systems and Seismic Countermeasures

for Field Facilities in NTT

by

Kazuhiko Fujihashi

1

ABSTRACT

NTT has experienced various disasters as a telecommunications operator, and has worked to develop a complete range of countermeasures.

Particular emphasis has been placed on

strengthening the early recovery systems when

disasters occur in order to improve the reliability

of telecommunications services. However, it was

extremely difficult to understand and convey

disaster conditions in the aftermath of the

Hyogo-ken Nanbu Earthquake, and this delayed

the start of disaster recovery activities. Drawing

upon this experience, NTT is developing and

introducing various disaster information

distribution systems and services, and is also

taking measures to improve the seismic

performance of the field facilities. Of these

efforts, this paper presents (1) support systems for

distributing and sharing information within NTT

and the NTT corporate group when a disaster

occurs, (2) various systems for estimating and

understanding the damage conditions of field

facilities, and (3) technology for improving the

seismic performance of field facilities.

KEYWORDS: Telecommunication Facilities,

Disaster Information System, Hyogo-ken Nanbu

Earthquake, Fiber Optic Sensing Technology

1. INTRODUCTION

The Hyogo-ken Nanbu Earthquake was a major

earthquake with local severe tremors of degree

VII (JMA seismic intensity scale), and produced

large-scale liquefaction in coastal areas and

reclaimed lands. For these reasons even NTT

telecommunication facilities were unable to remain unscathed. In addition, understanding of

disaster conditions and dissemination of this

information was slow in the initial stages, and

this delayed the start of disaster recovery

activities. At the same time, calls flooded the

circuits from all over the country to Kobe asking

for information, creating congestion never before

experienced and reaffirming once again the

importance of information during disasters.

Following the Hyogo-ken Nanbu Earthquake and

in consideration of the importance of telecommunication services in an information society,

NTT took the lessons of this earthquake to heart

and established the Committee for Severe

Metropolitan Disaster Countermeasures as an

in-house committee for reviewing both hardware

and software aspects in order to strengthen

disaster countermeasures. This committee is

further divided into 7 special committees that

have conducted wide ranging investigations 1) . This paper presents systems that have been

investigated and introduced in this manner, or which are currently under development. Table 1 shows the relationship between the basic items

investigated by these committees and the systems

described in this paper.

1 Senior Research Engineer, Supervisor, Disaster Prevention & Environmental Protection Group, Civil

Engineering Project, Access Network Service Systems Laboratories, Nippon Telegraph and

Telephone Corporation (NTT)

267
2. TELECOMMUNICATION SYSTEM

CONFIGURATIONS AND DISASTER

PREVENTION FEATURES2'

Fig. 1 shows an overview of the NTT

telecommunication facility configuration.

Telecommunication facilities are broadly divided

into switchboard system equipment (node system)

that process information and cable and wireless

faculties (link system) that transfer information. Also,

the telecommunication network is broadly divided

into sections between telephone office where

switchboards and transmission units are installed,

and sections from user residences to telephone office,

with the former called the trunk circuit network and

the latter called the access system network.

2.1 Node System Facilities

Node system facilities are mainly installed in

telephone office. In these cases the building itself has

extremely high seismic performance, and the

facilities are installed using methods that are both

strong and offer superior anti-vibration performance.

In addition, even should the facilities receive some

damage, reserve systems are provided to minimize

the effects. Furthermore, telephone calls and other

traffic are monitored and controlled online, allowing

automatic switching to the reserve system at the first

sign of trouble. Storage batteries and emergency

generators (some locations) are also installed as

countermeasures against possible commercial power

stoppages.

Although the Hyogo-ken Nanbu Earthquake caused

some loosening and breakage of the switchboard

foundation bolts and ceiling fixtures, there was no

particular effect on service. However, commercial

power stoppage, battery damage or discharge, and

reserve engine damage all occurred at the same time,

so the switchboards stopped and up to 285,000 lines were temporarily disabled and unable to make or

receive calls.

The biggest problem for node system facilities is

congestion so great that the number of inquiry calls exceeds the switchboard processing capacity,

rendering it impossible to make or receive calls. In

the aftermath of the Hyogo-ken Nanbu Earthquake,

the number of calls reached a maximum of

approximately 50 times the normal peak traffic,

making it difficult to connect for five days. This

congestion is a characteristic trouble for

telecommunications facilities that does not exist for

other lifelines. (Fig. 2)

2.2 Link System Facilities

2.2.1 Trunk circuit network

The trunk circuit network is comprised of wired

and/or wireless sections, and has a multi-route and

double-route configuration. Thus, should service be

interrupted over a section, the system still allows

traffic to flow by switching to the reserve system of a

different route. For example, NTT maintains

separate routes between Tokyo and Nagoya along

the Tokaido and Chuo Route in both wired and

wireless for a total of six routes. Tunnel networks

have also been constructed in the central areas of

Tokyo, Nagoya and Osaka to create a network with

high reliability as underground facilities.

2.2.2 Access system network

The access system network is comprised mainly of

metallic cables. Most of these are aerial cables, and

only 21% of the total are underground cables

installed in conduits and other underground facilities.

In addition, portable phone systems are spreading

rapidly in recent years, with most customers using a

combination of wired and wireless communications.

The access system network generally adopts a star

(spoke) configuration, and alternate routes are not

secured with the exception of certain regions or

important users. The access system network covers a wide area, and

268
includes vast amounts of facilities and diverse

facility environments. Urban areas have high

concentrations of older facilities, so localized

damage occurs easily and it is not easy to estimate

the damage conditions. Most disruption of service

when an earthquake occurs is the result of damage to

these access system facilities. The Hyogo-ken Nanbu Earthquake caused building

collapse and fires, which severed aerial and pull-in

cables. In terms of conduits, manholes, underground

cables and other underground facilities, underground

cables suffered damage due to conduit breakage and

disconnection and manhole duct sleeve damage, but

damage leading to the disruption of service was

much less than for aerial cables, thus confirming the

high reliability of these facilities.

3. DISASTER INFORMATION SYSTEM

CONCEPT

Possible disaster information system functions

include detection of disaster occurrence, estimation

and understanding of damage, support for recovery

activities, and so on. As mentioned above, the

greatest need in telecommunications is systems for

understanding disaster conditions and assisting early

recovery. Of these systems, online information

systems in particular use the telecommunications

network, and are differentiated from offline

information systems that gather information directly

from people at the disaster site through patrols, etc. An overview of the online and offline information

systems thought to be necessary for

telecommunications is described below.

3.1 Online Information Systems

The ideal online information system should

constantly monitor the network to make sure that

telecommunications service is provided properly,

and be able to understand trouble occurrence in real

time and perform network control to minimize any

drop in service. In addition, it is considered

important and effective to construct a system that

would allow centralized control and sharing of

damage and recovery information in order to promote a smooth recovery when a large-scale

disaster occurs. As mentioned above, access system

faculties are vast in number and span a wide area, so a system that could gather disaster information

remotely with as little human assistance as possible

would also be extremely effective for promoting

early recovery.

3.2 Offline Information Systems

When an earthquake occurs, it may be difficult to

understand the damage conditions to telecommunications facilities using only the

above-mentioned online information systems. In

addition, online information systems are basically

used to estimate and understand whether damage has

occurred, and it is difficult to understand the degree

of damage. Furthermore, the damage conditions of

conduits, utility poles, buildings and other structures must essentially be surveyed directly on-site by

patrols. Therefore, offline information systems that

use this type of human intervention are also

necessary for recovery. However, these activities

should proceed after careful selection of survey

methods and priorities based on the estimation

results of the online information system In addition,

systems and tools are also necessary for obtaining

plant records and other on-site information needed to

gather disaster and recovery information, and for

compiling information gathered on-site into

databases, etc.

4. CURRENT STATE OF DISASTER

INFORMATION SYSTEMS RELATED TO

SERVICE, OPERATIONS AND NETWORKS

4. 1 Network Monitoring and Control Systems

These systems were constructed by NTT prior to the

Hyogo-ken Nanbu Earthquake. The systems were

originally installed in prefectural units, but with the

restructuring of NTT in January 1999 they were

269
integrated into wide-area groups, with ten network

operation centers and two network control centers

currently located around the country (Fig. 3). The

network control centers monitor and control the

trunk circuit network in real time 24 hours per day

using the following systems. 1) Overall Traffic

Monitoring System which monitors and analyzes

communications conditions in full detail, 2) Traffic

Control System which detects and controls

switchboard volumes as necessary when tolerable

limits are exceeded, 3) Common channel signaling

Network Monitoring and Control System which

monitors and analyzes the signal network between

switchboards, and 4) New Synchronous Network

Control System which performs remote control of

transmission routes.

Furthermore, the network operation centers monitor

and control the vast amount of telecommunications

facilities including transmission facilities using the

following systems. 1) Switchboard Maintenance

Support System which monitors switchboard

operating conditions and performs remote control

and testing, 2) Control Measures Support System

which monitors the overall communications

facilities including switchboards, transmission units

and wireless units, and 3) Transmission Monitoring

Control System which monitors transmission unit

operating conditions, determines breakdown

locations, and switches to reserve transmission

routes.

4.2 Disaster Information Transmission and Sharing

Systems

33

4.2.1 Disaster information system

The disaster information system manages

information such as the activity items of each NTT

Disaster Countermeasures Headquarters team,

damage conditions and recovery conditions using

various databases when a disaster occurs. The

system distributes this information and also displays

it on a map. This client-server type system uses a LAN linking various corporate locations, and

performs centralized management from the Disaster

Countermeasures Headquarters server. The system

uses commercially available Lotus Notes software

that has been provided with group-ware functions.

When a disaster or accident occurs and disaster

information is input from a client, the disaster

location and area are displayed on a map. This

disaster information can be easily referenced by

simple point-and-click operations. Current facility

conditions and other data can also be displayed

during normal periods using the facility management

database of this system.

The functions of this system are: 1) map display of

disaster locations and areas, 2) display of disaster

countermeasures system information, 3) display of

various instruction and report information, 4) display

of facility disaster recovery information, and 5)

display of service call number "113" malfunction

reception information. This information can be

categorized and searched according to combinations

of time, source/location, information type and other

factors. In addition, various information required

when disasters occur such as "Disaster

Countermeasures Equipment List", "Disaster

Countermeasures Goods List" and "Emergency

Rations List" are compiled into databases.

4.2.2 Maintenance information network

The Maintenance Information NETwork (MI-NET)

is an electronic bulletin board where information can be input and viewed in text format. This system can

be accessed using web browser software from all PC

terminals of the four NTT companies (NTT, NTT

West, NTT East, NTT Communications) and

specified PC terminals at each NTT Group company.

This allows information sharing between all group

organizations.

In April 1999 the entire NTT Group conducted a

large-scale breakdown exercise that simulated a

leased line accident with the aim of verifying the

response during disasters. This system was used as

an information linking tool during this exercise, and

its practicality and effectiveness were confirmed. As

270
a result, while there was some need for enhancement

of functions, the system was confirmed to be

effective overall at speeding up information transfer

and information sharing.

4.2.3 Disaster Emergency Message Dial4)

Disaster Emergency Message Dial is an information

transmission system that makes full use of the

nationwide telecommunication network to prevent

the massive amount of communication following a

disaster from causing congestion. This is a voice

mail system for conveying personal safety and other

information verbally using phone numbers within

the disaster area as keys. To use this system, users

dial "171" and follow voice guidance to register and

playback messages. Unlike conventional voice mail,

message registration and playback are dispersed

throughout the country by distinguishing the

telephone numbers for registering messages using

the last three digits within the NTT network (Fig. 4).

The main uses are: 1) to contact people who cannot

answer calls due to evacuations, etc. and 2) to allow

contact when residential phones cannot be used due

to power failure or disasters. This system is expected

to prevent congestion and reduce the effects on

communication for disaster relief and recovery

activities.

Disaster Emergency Message Dial was operated

during the heavy rains and flooding that occurred in

Tochigi and Fukushima Prefectures in August 1998

and also during the Iwate-ken Shizukuishi

Earthquake that occurred one week thereafter, and

the system was accessed approximately 69,000

times including both registrations and playback. The

system has also been used a number of times since,

and is also operated on a trail basis during the annual

Disaster Prevention Week. Through these operations,

the service has been confirmed to be an effective

tool for transmitting information and solving the

problem of congestion when disasters occur.

Recently, the service was accessed 200,000 times

during the Tottori-ken Seibu Earthquake (October 6,

2000) and 87,000 times during the Geiyo

Earthquake (March 24, 2001).

4.2.4 Shared Map Database System0

This system consists of a shared map information

system for compiling map information databases

and various map utilization systems that use this

map information. The client-server system

configuration has two dispersed databases: a map

information management database and an attribute

information management database. NTT had

previously developed information systems that

utilize map data, but these databases generally could

be used only with the system in question. This

system resolves this issue, allowing databases to be

used by various systems. Map information position

coordinate data can be supplied from figure data or

attribute data, and figure data can be supplied in

various formats.

Examples of applying the Shared Map Database

System to disaster information systems include: 1)

linking mobile technology with map information, 2)

understanding damage conditions and searching

damaged facility drawings, 3) searching damaged

buildings and disaster prevention facilities, 4)

arranging recovery work. 5) supporting on-site

surveys, and 6) providing information on the web,

etc. The ability to search and display these and other

various types of information on maps is effective for

sharing information.

5. CURRENT STATE OF DISASTER

INFORMATION SYSTEMS RELATED TO

FIELD SYSTEM FACILITIES

5.1 Application for Evaluating the Seismic

Performance ofAccess System Underground

Routes^

5.1.1 Overview of the Seismic Performance

Evaluation AP

Damage to underground commumcations facilities

271
caused by the Hyogo-ken Nanbu Earthquake was

relatively minor. However, underground

communications facilities span vast areas, and visual

confirmation of damage conditions is difficult, so

facility inspections take much time which posed a

significant obstacle to recovery work. Drawing on

this experience, NTT developed the Application for

Evaluating the Seismic Performance of Access

System Underground Routes (hereafter "Seismic

Performance Evaluation AP"). This tool evaluates

the seismic performance of underground conduit

facilities based on earthquake, ground and facility

information, and is used to estimate macro damage

after earthquakes.

Fig. 5 shows an image of the AP. Advance

evaluations conducted during normal periods are

used to formulate overall plans and renewal plans,

and macro damage estimations after earthquakes are

used to calculate recovery costs and formulate

emergency inspection plans, etc. The databases

required by this AP are the various in-house shared

databases, which greatly reduces the database

compilation and maintenance costs.

5.1.2 General evaluation flow

Fig. 6 shows the general AP evaluation flow.

Analysis results of the emergency inspections of

underground conduit faculties conducted

immediately following the Hyogo-ken Nanbu

Earthquake showed that underground conduit

facihty damage factors were broadly classified into

liquefaction occurrence, the degree of seismic

motion, the conduit type, and the year of

construction. The AP can be broadly divided into a ground evaluation block and a facihty evaluation

block. The features of this AP evaluation flow are: 1)

comprehensive liquefaction judgment using the

micro-topography classification, boring data, and

liquefaction risk factor maps issued by each

municipality, and 2) ground evaluation in fixed

distribution (FD) sections which are the unit sections

used by NTT when managing the communications

network. At the initial settings, topographical data

and boring data are input and modified to evaluate

the ground quality. Next, earthquake information is

input and the seismic motion is estimated using a

distance attenuation formula. Then, the AP makes an

overall liquefaction judgment based on this seismic

motion and ground information.

Facihty evaluations target the six facihty types of

manholes, hand holes, building lead-in conduits,

facilities attached to bridge girders, main line

conduits, and distribution line conduits. Each facihty

is divided into a number of categories according to

the structure type and year of construction, and the

damage probability is estimated and evaluated

according to the combination of liquefaction

occurrence and the maximum ground surface

acceleration (PGA).

The estimation and evaluation results are displayed on a map by colors corresponding to the FD unit

liquefaction judgment results and the damage

probability for each facihty. Selecting a FD section

with the mouse opens a window displaying the FD

number, liquefaction judgment results, measured

seismic intensity and other data. Also, selecting a

span with the mouse opens a window displaying the

facihty name, facihty section, average, maximum

and minimum damage probabilities, number of

conduits and other information. Selecting the

[Detailed Conduit Information] button in this window opens a window displaying detailed

information such as the damage probability and type

of cables housed for each conduit. Fig. 7 shows an image of the evaluation results window display.

5.2 System for understanding Access System

Damage Conditions

In contrast to the AP described above which is a tool

for evaluating the seismic performance and

predicting damage to underground conduit facilities

that house cables, this system is a tool for testing

cables to predict and understand the effects on

service. This system is used for 1) estimation of

macro damage conditions in about half a day

immediately after a disaster occurs, and 2)

272
understanding damage conditions in the recovery

period. Fig. 8 shows an overview of the system.

5.2.1 Estimation of macro damage conditions by

order line tests

Communication cables include order lines for each

fixed distribution section that are lines used for

contact during work. When estimating macro

damage conditions immediately after a disaster, first

an order line G/NG judgment is made using order

line testing units, and these results are sent to and

registered in a database. Next, a macro estimation of

damage conditions is made using the off-site facility

management system (MARIOS), and these results

are indicated on maps and in ledgers.

5.2.2 Understanding damage conditions by all-line

tests

Disaster conditions in the recovery period are

understood by testing all lines from the MDF (main

distribution board) inside the facility center buildings

using automatic line switching and monitoring units.

The test results are sent to and registered on a server, and the damage conditions are indicated on maps

and in ledgers.

5.2.3 Other functions and features

In addition to macro estimation immediately after a

disaster and understanding damage conditions in the

recovery period, transmission technology that

supports multiple communication means such as

public network, leased lines and portable phones has

also been established to enable reliable data transfer even under disaster conditions.

5.3 System for Monitoring Structure Degradation

Using Fiber Optic Sensing Technology

NTT has researched and developed B-OTDR

(optical pulse tester using Brillouin scattered light)

based strain measurement technology as original

fiber optic sensing technology.

7' 8) NTT has applied

this technology to develop a system for monitoring

cracking and other degradation caused by

earthquakes, superannuation, adjacent construction,

etc., and has begun introduction to communications

tunnels in metropolitan areas.

5.3.1 Measurement principle

Fig. 9 shows the principle of fiber optic strain

measurement using the B-OTDR method. When

pulsed light is input to an optical fiber, Brillouin

scattered light having the frequency shifted in

proportion to the strain generated in the optical fiber can be observed. The position at which the strain is

occurring can be determined by measuring the time

for the scattered light to return. Therefore,

integrating structures and optical fiber makes it

possible to continuously measure the strain

generated in structures.

5.3.2 Overview of tunnel monitoring system

9)

NTT has developed and started introduction of a

system for constant remote monitoring of the strain produced in each section when cracking or other

degradation occurs. This system uses optic fiber cables

fixed at set intervals to the inner walls of

communication tunnels as strain sensors. Fig. 10 shows

the system configuration, and Table 2 gives the

performance. Tunnels within a radius of approximately

10 km can be monitored by a single B-OTDR using

optical switches, and cracking on the concrete surface

can be detected in 0.1 mm units for a fiber fixing

interval of 2 m. This system makes it possible to

quickly determine the presence of degraded locations

after a major earthquake, and also monitors concrete

cracking and peeling due to superannuation and other

factors during normal periods. This is thought to help

reduce maintenance costs by enabling quick detection

and repair.

5.3.3 Development of Optical Fiber Sensing

Technology to a Wide-area Disaster Monitoring

273
System

In addition to monitoring tunnel degradation, NTT is

also working to apply fiber optic sensing technology

to other R&D projects such as monitoring road

escarpment collapse and river embankment collapse.

B-OTDR fiber optic sensing features the ability to

continuously measure the longitudinal direction for

long distances, and development to applications such

as monitoring changes in large structures that cover a wide area and ground conditions can be expected in

the future.

6. SEISMIC COUNTERMEASURES FOR FIELD

FACILITIES

The Hyogo-ken Nanbu Earthquake also caused great

damage to telecommunications facilities. Comparing

communication cable damage conditions for

underground and aerial facilities shows a damage

ratio of 1 to 9 at the service level and 1 to 30 at the

facility level. These figures confirmed once again

that burying communication cables is an effective

means of ensuring communications reliability.

However, weak points were also confirmed such as

the connections between different structural systems,

so technology was developed and introduced to improve the seismic performance of these

portions.

10

* Fig. 11 shows an overview of these

technologies. Priority issues were the three locations

of expandable joints for open-cut tunnels, manhole

ducts, and building lead-in portions. In addition,

cracking and protrusion also occurred at the

connections between shield tunnels and pits due to

the lack of reinforcing bars, so expandable joints

with superior waterproof performance were recently

developed.

6.1 Flexible Tunnel Joints

The Hyogo-ken Nanbu Earthquake caused no

damage whatsoever to cables housed in tunnels,

indicating that the tunnels fulfilled their intended

function to protect the cables. However, of the

expandable joints of open-cut tunnels installed in the

liquefied ground, grade differences of up to 18 cm in

both the vertical and horizontal planes occurred at

the connections between the tunnels and buildings or

pits, resulting in leakage and flooding. Therefore,

flexible joints were developed as seismic

countermeasures for the joints of open-cut tunnels in

ground subject to liquefaction. Workability was

improved to allow execution with as little movement

of cables as possible, and these flexible joints have

been installed in 130 locations thus far. In addition,

cracking and leakage were also found in three

locations where shield tunnels connect to pits, so

joints capable of withstanding water pressure up to

0.7 MPa and having expandable performance of ±10

cm in the direction of the tunnel axis were also

developed.10 (Photo. 1)

6.2 SFRC Manhole Ducts

Manhole ducts are the connection points between

the different structural systems of manholes and

conduits, so ground deformation and other factors

during earthquakes cause relative displacement

between the two, applying large force to these

portions. In areas with significant damage, there were cases where the damage did not stop at

cracking, and concrete lumps also peeled away and

damaged the communication cables. Therefore, steel

fiber reinforced concrete (SFRC) was adopted as a method of providing the duct concrete with

sufficient load bearing strength after cracking occurs.

6.3 Flexible Building Access Conduits

When ground subsidence and other factors during

earthquakes caused comparatively large relative

displacement between hand holes and customer

buildings, the building access conduits were unable

to absorb this variation and suffered damage such as

conduit separation and breakage. As a

countermeasure, a flexible joint structure capable of

absorbing large displacement at the building end and

joint portions was adopted.

274
7. CONCLUSION

Based on the experience of the Hyogo-ken Nanbu

Earthquake and in addition to the various

countermeasures for large-scale disasters that have

been adopted thus far, NTT is pursuing new

countermeasures aimed at realizing disaster-proof

communications services. As the importance of data

communications increases in the future, NTT intends

to construct even more reliable networks and

develop various systems in order to fulfill its social

obligations in times of emergency.

REFERENCES:

1)Honda, et al.: Telecommunication Facilities, Damage

and Restoration of Lifeline Facilities, Hanshin-Awaji

Great Earthquake Disaster Investigation Report

vol.9pp.467-505, 1997(in Japanese)

2)Fujihashi , Komatsu,: Real-time Disaster Information

Systems Related to Telecommunication, Proceedings of

2

nd Real-time Earthquake Disaster Prevention

Symposium, JSCE, pp. 1 35-142,2000(in Japanese)

3)Fujihashi,Shigeta : Systems related to disaster

countermeasure in NTT,UJNR Earthquake information

system workshop,2000

4)Hashimoto, et al.: Disaster Emergency Message Dial

Starts Operation, Journal of NTT Technology, 1999,

Vol.10 No.3 (in Japanese)

5)Nakagawa et al.,: Disaster Measure Support Systems

Using Common Map Information Database, Journal of

NTT Technology, 1 999 , Vol. 1 1 No3(in Japanese)

6) Honda et al.: Development of Earthquake-proof

Performance Evaluation Application for Underground

Telecommunication ConduitsN 3rd China-Japan-U.S.

Lifeline Earthquake Engineering Symposium^ 1 998

7)Kurashima, et al :Optical fiber sensor for distributed

strain measurement in concrete structures, 19* Meeting

on Lightwave Sensing Technology, 1 997

8) NTT News Release, Sep.2001

,

URL:http://www.ntLco.ipAiewsAiewsO 1 e/01 09/0 1 0920.ht

ml

9)Takatsuka, Fujihashi, Telephone Tunnel Monitoring

System Using Fiber Optic Sensing Technology,56lh

AnnualConerenceJSCE,2001 (in Japanese)

10)Honda et al. :Aseismic Countermeasure Technologies

for Outside Telecommunication Facilities

Implemented Following the 1995 Hyogo-ken Nanbu

Earthquake^ Draft Proceedings of the 7th US-Japan

Workshop on Earthquake Disaster Prevention for

Lifeline SystemsN 1997

11)Okutsu,Fujihashi, :Study of Flexible Joints installed at

the Connections between Shield tunnels and Pits, 57

th

Annual ConferenceJSCE,2002(in Japanese)

275
Table 1 Contents Investigated by the Committee for Severe Metropolitan Disaster Countermeasures and Disaster

Prevention Related Systems.

ITEM CONTENT SYSTEM

Securing and control of

telecommunication

resource

OSecuring of communication resource by destruction

prevention and up-grading the reliability of

communication equipment

OReinforcement ofnetwork control function.

ODestruction prevention in network and reinforcement

of control function

Allotment and distribution

of telecommunication

resource

OAllotment, distribution and the operation of

telecommunication resource when crowded or

destructed.

Disaster Emergency Message Dial

Disaster area Information

network

OConstruction of a new information circulation

mechanism in disaster area, which uses die ISDN line,

personal computer, LAN, etc.. Crisis-management/

restoration formation

OMaintenance of crisis-management system

OQuick ascertainment of damaged area within the access network.

Disaster Information System

Maintenance Information NW

Access System Underground

Routes Seismic Performance

Evaluation AP

Access System Damage Condition

Understanding System

Shared Map Information DB

Fiber Optic Sensing Technology

Fig. 1 Telecommunications Facility Fig. 2 Communications Congestion

Configuration Overview Immediately Following

the Hyogo-ken Nanbu Earthquake

276
O Nationwide NW Control Center o Regional NW Operation Center

Fig. 3 NTT Network Monitoring and Operation

System

Other Area

I 'Message Accumulation Device O50 devices are distributed to major cities. OMessages ate allotted by treble of the telephone number. OAble to accumulate 8 million messages. y

: Message Recording

% : Message Playback

Fig. 4 Overview of Disaster Emergency Message

Dial

Cdor-coded dsplay of facility damage probability and

liquefaction judgment

Fig. 5 Image ofApplication for Evaluating the

Seismic Performance ofAccess System

Underground Routes

Initial Settings

Ground

Evaluation Part

Fsdilty Evaluation

Part

Input of

terrain data

i

Extraction and

settings of boring data

Inpjjt of aartJ

-

almuta

(Munfclp

quake etiaaatar Honrwul

aHDaa data)

: ——i

inpu* of E*p»cta<5 Earthquake information

Earthquake Scale Esttmatk*

£80 gal

Damage probability judgment

Display

Mo damages

Rough estimation

of restoration cost

Display (maps, tables, text etc.)

Fig. 6 General Evaluation Flow of Seismic

Performance Evaluation AP

Fig. 7 Image of Evaluation Results Window Display

277
Table 2 Tunnel Monitoring System Performances

Damage MARIOS

Information Server V Server J

Fig. 8 Overview of System for Understanding Access

System Damage Conditions

ITEM System using

B-OTDR

Present Systems

usine Electric

Sensors

Measurement

Object

Measurement of

continuous section

Measurement at point

Measurement

Distance

Up to JUkm T T_ O.-.. .it,.!.-. 1 1 f\f\., . Up to Several luum

Strain

Measurement

Accuracy

50^

Ability to detect crack of0 1mm

within 2m section

fewjiz

Electric

Power Supply

Unnecessary.

Thunder

countermeasures

are unnecessary.

Necessary for each

sensor. Require thunder

countermeasures. Sensor

Wiring/

Distribution

Single optical

fiber. Regardless of

measurement

point

Each sensor needs

leading line. Becomes complex as measurement point

increases.

Fig. 10 Tunnel Monitoring System Configuration

278
Fig.l 1 Overview of Seismic Countermeasures for Underground Structures

Following the Hyogo-ken Nanbu Earthquake

279

SPECIAL THEME 3

MANAGEMENT OF

INFRASTRUCTURE SYSTEMS

Management of Infrastructure Facilities as Disaster Protection

by

Tadayuki Tazaki\Takeo Nakajima2

, Kazuhiro Nishikawa3

ABSTRACT

NILIM was established in April, 2001 following

the reform of the central government.

NILIM's organization and research policy was

introduced. Its three comprehensive research

centers characterize NILIM.

This paper concludes that discussing subjects,

such as Management of infrastructure facilities

to ensure the function as disaster protection,

Disaster information technology and disaster

mitigation, Application of ITS technologies as powerful tools for mitigating disaster, will be an

important role for the UJNR joint panel meeting

in the 21st century.

KEYWORDS : NILIM

Technological Policy

Management of infrastructure

Disaster Information

Application of ITS technology

1. INTRODUCTION OF NILIM

Japanese central government was restructured as of January 6, 2001. The number of ministries

was drastically reduced to one cabinet office and

12 ministries. The Ministry of Land,

Infrastructure and Transport, or "MLIT", was

formed as a combination of the functions of

these four ministries and agencies.

There were three research institutes relevant to

construction in MLIT. These institutes were

separated from the ministry and in principle

transferred into Independent Administrative

Institutions.

The functions of these three original institutes

were taken over by these new institutions. Each

of the three original institutes functioned to support the headquarters to make technological

policies. NILIM has been established to take over half of researchers and this function.

When we are asked to explain the differences

between NILIM and the three agencies. We

reply that NILIM is for management, and

considers the infrastructure requirements,

whereas agencies are for technologies, and

consider the performance of facilities.

For this purpose, NILIM was established in

April, 2001 as the only national research

institute in the field of infrastructure to support

the headquarters of MLIT.

The figure shows the organization of NILIM.

There are ten research departments and three

comprehensive research centers. It could be said

that these three comprehensive research centers

characterize NILIM. They are organized based

on demand-oriented subjects. They are engaged

in cross sectional research activities.

They are the Research Center for Land and

Construction Management, Research Center for

Advanced Information Technology, and

Research Center for Disaster Risk Management.

The last one deals with the subjects closest to

this joint panel, but the others also have relevant

studies.

1) Director General, National Institute for Land and Infrastructure Management (NILIM), Ministry of

Land, Infrastructure and Transport (MLIT)

2)Director of Planning and Research Administration Department, NILIM, MLIT

3)Research Coordinator for Evaluation, NILIM, MLIT

283
2. THE RESEARCH POLICY OF NILIM

The first four months was spent drafting the

Research Policy of NILIM, because NILIM

gathered researchers from three different

institutes and had to show the direction in which

to proceed together. This was a very important

and basic process at the beginning of the new

organization management.

The research policy was settled on July 30,

2001, and consists of Mission, Research

Subjects and Research Activities.

The Mission says "We conduct research and

development to contribute to technological

policy-making that enhances the satisfaction of

citizens as the end-users of infrastructures".

I would like to emphasize the necessity of taking

into account the demand-side point of view in

the research activities.

The goals are stated as:

- Conservation and creation of desirable national

land

- Guarantee of safe and secure life

- Enhancement of quality of life

- Vitalization of civil society and economy

To achieve the goals, seven principal subjects

are selected. The first four subjects are relevant

to the goals of MLIT.

1 . Creation of sustainable society with a

beautiful national land

2. Building of a safe and reliable national land

3. Enhancement of quality of life

4. Vitalization of civil society and economy

We believe we can construct an image of Japan

in the 21st century through research on these

four subjects.

The last three subjects are relevant to the

procedures for MLIT's activities.

5. Improvement of construction management for

public works

6. Establishment of foundation for advanced

information society

7. Contribution to international society

Under these seven principal subjects, sixteen

technological research subjects are provided in

the policy.

Here are some examples:

- Asset management

- Disaster information technology

- Intelligent transportation system

- Public involvement

The first three subjects will be mentioned later. The last one, study on public involvement, is now indispensable regarding public works

procedures, of course, including disaster

prevention and mitigation. Nowadays it is impossible to carry out public works without the

participation of citizens.

3. ASSET MANAGEMENT

Introducing the research subjects along with the

three research centers will help to understand the

institute more deeply and what kind of subjects

we propose dealing with at the joint panel

meeting.

The first one is the Research center for land and

construction management.

There are three subjects.

The first one is "Historical study on national

land and infrastructure". It is important and

useful to study the past to find out the future

direction.

The second one is "Construction management

for public works". Effective implementation of

public works might be our permanent task.

The third one is "Asset management". We often use the term "Stock Management" instead.

284
Asset management is very important from

various aspects. These include: dwindling

availability of disposal sites and increasing C02

emissions regarding the global environment,

both of which are exacerbated by the renewal of

infrastructure. Furthermore, massive demand for

renewal will place a big burden on the

decreasing younger generation. It is therefore

essential to prolong the life of infrastructure

with efficient management.

Aging and deterioration are inevitable for every

infrastructure facility, but this also reduces the

capacity against natural disasters.

To overcome this and to create a sustainable

society, strategic management as a combination

of inspection, diagnosis, rating, repair,

strengthening and replacement technologies are

necessary.

4. DISASTER INFORMATION

TECHNOLOGY AND DISASTER

MITIGATION

The next is the Research center for disaster risk

management.

Its major subjects recently are Disaster

information technology and disaster mitigation.

Research on disaster prevention is now

spreading to include the use of information from

satellites.

Hazard maps for various kinds of disaster and

damage prediction technology provide valuable

information.

Gathering damage information by mobile tools,

integrating them with GIS, and distributing

necessary information to people are extremely

effective disaster mitigation methods.

5. APPLICATION OF ITS TECHNOLOGIES

The last one is the Research center for advanced

information technology.

Its main research concerns ITS (Intelligent

transport system), which seems to have little to

do with disaster prevention or mitigation. But

the application of developed technologies using

new information technologies can be powerful

tools for mitigating disaster.

For example, a car navigation system, already

installed in over 10% of automobiles in Japan, can guide drivers along detour routes when they

come across major damage caused by natural

disaster. And ITS for pedestrians, developed

originally for handicapped people, can inform

strangers by cellular phone of where to evacuate

in case of emergency.

6. CONCLUSIONS

UJNR has a tremendous track record from the

20th century. Many new technologies have been

adapted in the design codes in both countries

through the UJNR research-and-development

programs in the field of earthquake disaster

prevention and wind effect technology. The 21st

century has just started, and we must create a

sustainable society for future generations.

There are three subjects here that I have already

mentioned. Discussing these kinds of subject

will be an important role for the UJNR joint

panel meeting in the 21st century. We are standing at the crossroads of the new

millennium. We must define a suitable direction

for UJNR activities and exchange the latest

information each other.

285
Management of Infrastructure

Facilities

to Ensure the Function

as Disaster Protection

Tadayuki TAZAKI, Takeo NAKAJIMA

Kazuhiro NISHIKAWA

National Institute for Land and

Infrastructure Management

Central Government Reform

1 Office and 22 Ministries

Reformed jg January 6, 2001

1 Cabinet Office and 12 Ministries

Ministry of Construction

Ministry of Transport

National Land Agency

Hokkaido Development Agency

Ministry of Land,

Infrastructure and

Transport

MLIT

286
Independent Administrative Institution?

National Institute

Former PWRI

April 1, 2001

Transformed

Independent Administrative Institution

New PWRI

NILIM

&

Independent Administrative Institution

Public Works

Research Institute

Building

Research Institute

Port and Harbour

Research Institute

Reformed

April 1, 2001

NILIM

251 researchers

3 Independent Administrative Institutions

287
What's the difference?

NILIM vs. IAIs

(^Management C^Technology^^>

Infrastructure requirements Performance of facilities

Establishment of NILIM

The only national institute in the field of

Land and Infrastructure Management

search and develop

support technologi

policy-making of ML

April 1,2001

288
Organization

Director-General — 2 Deputy Director-Generals — Executive Director for Research Affairs

General Affairs Department 3 Administration depts.

Planning and Research Administration Department

Administrative Coordination Department

Road Department

River Department 10 Research depts.

3 Comprehensive research centers

Research Center for Land and Construction Management

Research Center for Advanced Information Technology

Research Center for Disaster Risk Management

Three Research Centers

Research Center for

Land and Construction Management

Research Center for

Advanced Information Technology

Research Center for

Disaster Risk Management

289
Research Policy of NILIM

July 30, 2001

Mission

Research Subjects

Research Activities

Mission

'We conduct R&D to contribute

to technological policy-making

that enhances the satisfaction of citizens

as the end-users of infrastructure.

"

290
Goals

SConservation and creation of

desirable national land

SGuarantee of safe and secure life

SEnhancement of quality of life

SVitalization of civil society and economy

Seven principal subjects

1. Creation of sustainable society with a

beautiful national land

2. Building of a safe and reliable national

land

3. Enhancement of quality of life

4. Vitalization of civil society and economy

291
Seven principal subjects

(cont'd)

5. Improvement of construction

management for public works

6. Establishment of foundation for

advanced information society

7. Contribution to international

society

Research subjects

(examples)

SAsset (stock) management

SDisaster information technology

SIntelligent Transport System

SPublic involvement

292
Research Center for

Land and Construction Management

SHistorical study on national land and infrastructure

sConstruction management for public works

SAsset (stock) management

(from "scrap & build" to "stock & renovation")

Stock Management to Ensure

Disaster-Protection Function

Aging and Degradation

Inspection, Diagnosis, Repair, Rating

Strengthening, Replacement

of

Existing Facilities

Strategic Management Is Required 11

293
Research Center for

Disaster Risk Management

Earthquake damage prediction system Common Information

Disaster information technology

and

Disaster mitigation

• Hazard maps

• Damage prediction

• Information

gathering by mobile tools

integrating with GIS

distributing to people

294
Research Center for

Advanced Information Technology

SPromotion of intelligent technology in transportation

GIS

Application of ITS Technologies

Distribution of Disaster Information

through

SCar Navigation Systems

to guide along the detour route,

SITS for Pedestrians (cellular phone)

for strangers to evacuate

295
Conclusions

The roles of the Joint Meeting in the 21st century

STo renovate and make good use of

existing infrastructures

STo promote the disaster fnformation

system with IT such as G/S, GPS,

mobile tools, etc.

STo apply ITS technologies as

disaster information tools

Information Available

Please visit us at:

'o "//www- nil i in iiojoi

296
Information Systems for Infrastructure Management and Emergency Response

by

William E. Roper

ABSTRACT

Information systems and their application to infrastructure management for emergency

response with particular focus on critical

transportation infrastructure are addressed in this

paper. The opportunities for information systems

leveraging geospatial data sources are presented.

Remote sensing imaging and non-imaging

sources of information are explored.

Coordination and sharing of in information is explored to improve the ability of multiple

organizations to more effectively manage and

respond to emergency situations. The use of

geospatial information was only a part of the total

recovery activity but did provide important

assistance throughout the recovery operation.

Samples of geospatial technologies used in the

recovery are presented. These include 3-D

visualization, thermal infrared imagery, LIDAR

data systems, IKONOS one-meter panchromatic

imagery, SPOT imagery and the use of digital

aerial imagery. The general area of disaster

response management is also addressed and the

findings of various studies in this area are summarized. Recommendations for the use of

information systems and geospatial sensors and

products in the future for infrastructure

management are presented.

KEYWORDS: Information systems, Critical

transportation infrastructure protection,

Geospatial information, IKONOS satellite data,

Digital imagery, Data management, Disaster

management, LIDAR, Infrared remote sensing,

GIS mapping, Global Positioning System

'Professor and Chairman Civil and Environmental

Engineering Department, Suite 643, Academic Center,

George Washington University, Washington, DC, USA,

20052 wroper@seas.gwu.edu; phone 202 994-4901 ; fax 202

994-0127; http://www.seas.gwu.edu;

1.0 INTRODUCTION

Infrastructure managers at all levels face

unprecedented challenges today. Increasing

demands are being placed on agencies to preserve

the existing infrastructure system and to take on

new missions of improved system safety and

security. A variety of advanced technologies are available to enhance planning, designing,

managing, operating, and maintaining all aspects

of the nations public and private facilities. The

national transportation system is one example of

a major part of our infrastructure system and a

critical asset for national growth and

development. Aerial and satellite remote sensing

represents one area of rapid development that can be leveraged to address these challenges. These

technologies have significant and unique

potential for application to a number of cross

cutting transportation security issues. A

summary of how some of these information

systems were used in the recovery effort in New

York City will be presented and assessed.

2. 0 INFORMATION MANAGEMENT

CONSIDERATIONS FOR EMERGENCIES

It is clear that despite excellent efforts by many

groups the approach to providing information for

disaster management is not often effectively

utilizing a wealth of data that resides, with

various organizations. This existing information

and technology could provide disaster mangers

important information products that could save

lives, reduce damage to property, and lessen the

environmental impacts of disasters. The current

situation is characterized by numerous

shortcomings that inhibit optimal decision

making for disaster management. The inability to access information and the lack of

standardization, coordination, and

communication are all obstacles that need to be overcome.

297
2.1 Information Management Goals

Considerable effort has been expended

throughout the Disaster Management community

to articulate issues and to characterize the

dynamics and the inter-relationships that needs to be accommodated in a viable information

management strategy. An important guide for

this effort is found in the recommendations made

by the Board On Natural Disasters (BOND) in

their report to the National Research Council.

The BOND'S primary goals as articulated in its cha were to:

1. Improve decision making before, during, and

after emergencies through better access to and

quality of data and information

2. Identify users and their needs

3. Provide information products specifically

designed to meet users' need

4. Promote efficiency and cost effectiveness

5. Stimulate and facilitate mitigation

These are important components in any workable

framework for comprehensive disaster

management. However, the focus here is directed

towards derived products, procedure and

protocol. It is important that the process develop

baseline products and implement solutions that

reflect the complex fabric of the disaster

management community. This work is done at a

lower, more formative level in the information

management life cycle and in fact forms the basis

for the production of the products referred to by

the BOND. As noted above the assessment

process will work to identify the constituent

components associated with specific disaster

events and scenarios. The constituent

components will then become the basis for

technology specifications and the delivery of

capabilities that encompass the products

identified by the BOND.1

2.2 Data Management Considerations

from recent studies in disaster management

community itself. These assumptions certainly

do not constitute and exhaustive set of

parameters. Rather, they provide a point of

departure and a set of "operating hypotheses" that can be used to help frame discussion and

consideration of needs and requirements.

Available Information and capacity is not

uniform, consistent and standardized for disaster

managers. Large cities may have invested

millions of dollars in their information systems,

while smaller municipalities may be operating

with one personal computer. Disasters will

constantly crosscut established boundaries. A

fundamental problem in dealing with disasters is

that they do not respect boundaries that include

organizational, political, geographic,

professional, topical, or sociological

consideration.

2

Disasters will often overwhelm mechanisms for

coordination and communication. In addition,

the mechanisms to bring data and information to decision makers are uncoordinated. Information

is often produced from disparate sources and

transmitted in whatever format the provider

prefers, requiring significant effort to compile it into a form that provides a coherent picture or even thwarting integration altogether. Data

standards are often inconsistent, and, even more

dangerous; users are sometimes unaware of the

limitations and uncertainties in data or are presented with conflicting interpretations of data

without the means to assess the reliability of the sources. All of these issues reduce the efficacy of

the decision-making process. The problem is compounded because information delivery

systems in many cases become overloaded during

crises. For instance, in 1996 a moderate

(magnitude 4.7) earthquake in San Jose,

California, led to more than a million attempted

hits in less than one day on the U.S. Geological

Survey's World Wide Web earthquake

information server.

3 Most of those attempts were

This section provides assumptions based on 2 Jobin, Graham A, and Burrell E. Montz, Natural Hazards:

information and conclusions that have emerged Explanation and Integration, New York, NY: The Guilford Press, 1997

3 Burton, Ian, Robert W. Kates, and Gilbert F. White, The

1 NRC Report, Information Infrastructure for Managing Nat\ira\Environment as Hazard, 2d Edition, New York, NY: The Guilford

Disasters, Board on Natural Disasters, National Academies Press,Press, 1993

1998

298
unsuccessful, including those by emergency

managers trying to access data and information to aid in the formulation of a response plan.

3.0 INFORMATION MANAGEMENT

PROCESS ISSUES

Disaster management is an exercise in logistics

and information processing and distribution. To

effectively undertake these tasks requires a good

understanding of disaster information

requirements and the characteristics associated

with disaster events knowing that each event is uniquely different. Disasters come in different

sizes, have different behavior and can be

categorized on the basis of their impact on natural

resources, agriculture, communities etc. They

can also be discriminated and classified along a number of dimensions including impact, severity,

duration, geographic setting and advance warning

(NAPA 1998). In order to develop the

Information Technology architecture for case

scenarios application it is essential to understand

the disaster event from the perspective of those

responsible for assimilating the data. The

effective use of this information for producing

operational plans to deal with the disaster event

and its aftermath is critical.

3.1 Consideration of Infrastructure

Information Phase Management

For instance, it is important to appreciate that the

four basic phases of disaster management may

actually manifest themselves in different ways

depending on the nature of the disaster. The

mitigation phase is essentially a continuous

process of preparation, analysis of performance

and subsequent modification and refinement. As

a result it will likely exhibit the smallest amount

of variance. However, the other phases

(preparedness, response and recovery) will likely

have significantly different attributes depending

on the nature of the event. For example consider

the difference in the preparedness phase that

would be associated with major seasonal flooding

along the Mississippi or the Missouri Rivers, the

Chaparral fires along the Pacific Coast, a Tsunami that impacts much of the Pacific Rim

and an earthquake such as Northridge or Loma

Prieta. In each case the geographic coverage and

the amount of time for planning and staging is very different. As a result, many of the other

systemic aspects of the disaster management

initiative will be different as well.

An event warning which impacts the manner in

which the preparedness phase is managed also

has implications and consequences for the

recovery phase (e.g. In the case of the Loma

Prieta earthquake the emergency response

resources were directed to the bay area despite

the fact that Santa Cruz and Watsonville had

more significant damage). Partially as a result of

this California is building the Tri-Net seismic

sensing network to measure shake and model

damage in order to make more informed

decisions regarding staging recovery resources.

3.2 Infrastructure Disaster Management

Communities

The disaster management is extremely

heterogeneous. In order to develop effective

systemic solutions to the needs of this community

any framework must address the need to effectively and efficiently share information

across institutional boundaries within and

between the groups that make up this community.

In essence an interface must be specified at each

boundary that separates levels in an agency or at the juncture between different agencies or between different sectors. These interfaces each

have at least three components: 1) technical, 2)

institutional and 3) institutional administrative.

The importance of these interfaces is also

proportional to the difficulty of implementing

them in an automated context. For instance, the

lightest weight interfaces should be between

levels within the same entity within a single

sector. In contrast the heaviest interfaces would

be those that attempt to span sectors.

4

A number of these communities have made

significant institutional investments in the

construction of foundation data products that feed

modeling, analysis and decision support in a range of contexts. Not surprisingly the situation

4 NRC Report, Information Infrastructure for Managing Natural

Disasters, Board on Natural Disasters, National Academies Press, 1998

299
in the disaster management community in regard

to data parallels spatial data issues across the

federal government and throughout state and

local government and the private sector. The

issue of data sharing by the disaster management

community was also addressed by the BOND.

Their findings recognized that despite the

importance of significant databases, their utility is impaired by a host of problems deriving from

incompatible formats, inconsistent geographic

reference systems, conflicting standards, and

other human-caused factors. Many of these

problems could be resolved and the value and

utility of the databases for disaster decision

making greatly enhanced through improved

organizational and technological coordination

with only an incremental increase in cost.

5 This

is clearly in the public interest to accomplish this.

Disaster management scales for international

agencies and governments to the individual.

However, a very significant level of

responsibility is vested at the local level. For

instance, it is the on the ground officials, many of

whom are local, that ultimately play the role of

first responders (e.g. police, fire, medical and

public works) and who also manage many of the

recovery operations. To be effective in this

context protocols must define communications,

database structure, data formatting, hardware and

software requirements, networking, quality

control, and other issues needed to assure the

linkage will improve decision making before,

during, and after emergencies through improved

access to quality information. In addition, such a linkage will provide information products that are

specifically designed to meet the needs of users

3.3 Common Characteristics of Disasters

Disaster management is not a linear process that can be documented easily in a flow chart with a

readily apparent beginning and absolute end

point. Rather, it is a cyclical process of

approximation, response and re-calibration that

involves many different actors whose roles in

relation to one another are likely to change based „ on circumstances and the stage in the process.

The one constant evident in the process is the

chaos that drives the system. Another absolute is that the effective, efficient application of

information technologies and products has the

capacity to improve the system in a number of

ways that will combine to save lives, mitigate

overall damage, conserve resources and

ameliorate human suffering.

To develop effective architectures and

technologies that meet the needs of the disaster

management community there must be a precise

understanding of the disaster management life- cycle (Mitigation, Preparedness, Response and

Recovery), the Information Communities that

combine to define the disaster management

community and the information processing

requirements associated with the cycle of data

development, dissemination, analysis and review.

In addition, and perhaps more importantly, there

must be a precise understanding of the dynamics

between these components and the "interfaces"

that these dynamics imply. Only with such an

understanding can we effectively model the

process and derive technology solutions that map

well into the business model of disaster

management.

6

3.4 Data and Information Processing

Requirements

Data needs, the characteristics of data and the

ability of data to contribute to analysis, response

and recovery is well understood. The basic

attributes of information that are considered most

important by this community include:

1. Timeliness: delivery of data and information in

time to drive decision making

2. Consistency: delivery of data and information

in a consistent, uniform manner

3. Understandability: delivery of data and

information in a manner that is appropriate and

understandable in the target community

NAPA Report, Geographic Information for the 21s' Century:

Building a Strategy for the Nation, National Academy of Public

Administration Press, 1998

6 Disaster Information Task Force Report, Harnessing

Information and Technology for Disaster Management, Global

Disaster Information Network, 1 997

300
4. Accuracy: precision in measurement and

observation

5. Flexibility: adaptability to multiple situations

These information attributes have evolved over

time within the disaster management community

representing the viewpoint of a range of

providers, users and consumers scaled from the

local to national level. In this regard the

communities involved transcend boundaries

between recognized sectors at each of these levels

(private, public, not-for-profit, NGO etc.).

4.0 TRANSPORTATION

INFRASTRUCTURE INFORMATION

SYSTEMS

The nation's transportation networks function as the lifelines of commerce, security, and

recreation. They are also vulnerable to attack and

severe disruption. As officials have discovered

since September 1 1 , protecting and preserving the

extensive U.S. transportation networks against

terrorist attacks remains a daunting task.

Fortunately, today's powerful geospatial tools,

especially remote sensing, positioning,

navigation, and timing (PNT), and geographic

information systems (GIS), can assist in this

crucial effort.

4.1 Overview of America's

Transportation Infrastructure

America's transportation infrastructure is a

reflection of the nation's economic vitality, which

in turn is predicated on maintaining stable

economic, political, and social environments.

Under most circumstances the nation's

transportation system works smoothly and

efficiently; but, as President Eisenhower

recognized in authorizing the Interstate Highway

system during the Cold War 1950s, it is vulnerable to attack from unexpected quarters.

Although the U.S. public had become

accustomed to peaceful stability within U.S.

borders, the appalling events of September 11,

2001, require a shift in thinking. While the

transportation system cannot be re-engineered in

its entirety to cope with this new operational

environment, a variety of coping strategies can

aid in protecting its many modalities and

components. Among these is a higher use of

aerial and satellite imagery (and collateral sensor

data from non-imaging systems), to obtain

intelligence for situations that develop rapidly

and which demand real-time decision-making.

The transportation system is extensive, complex

and critical to maintaining the U.S. economy

(table 1). The Administration's report, Securing

the Homeland, Strengthening the Nation , underscores the urgency in strengthening

transportation security. The proposed 2003 White

House budget includes not only the establishment

of a Transportation Security Administration

within the Department of Transportation, but

requests $4.8 billion "...to fulfill the mandates

established by the [Aviation and Transportation

Security Act of 2001]."

4.2 Potential of Image & Sensor Data for

Improving Lifeline Security

Remote sensing technologies, together with other

geospatial technologies such as GIS and PNT

systems, have a significant role to play in the

improvement of the Nation's transportation

security and critical infrastructure protection.

Remote sensing from space combines a broad

synoptic view with the ability to detect changes

in surface features quickly and routinely. Remote

sensing from aircraft provides the ability to examine areas in great detail from below the

clouds, and ground-based systems make possible

the close-in observation of events in real time.

Each of these spheres of remote sensing

technology hold benefits and drawbacks for use

in transportation, and in particular, for the

security of the Nation's complex transportation

networks.

8

Examples of the use of remotely sensed data to support transportation security include the ability

to:

See http://www.whitehouse.gOv/homeland/Q or http://whitehouse.gov/homeland security_book,html . Roper, William, Geospatial Technology Support to the Nation 's Navigation System, Transportation Research Board, National

Research Council, Washington, DC, January 1 999

301
• Develop accurate digital terrain models and

3-D surface features as a means for modeling

landforms along rights-of-way;

• Visualize terrain from different perspectives,

with the potential for developing threat cones and

viewsheds;

• Classify vegetation types along

transportation lifelines as a possible deterrent to concealment

;

• Detect, classify, and analyze temporal and

spatial changes in surface features; • Identify facilities where topography or

identifiable hazards (e.g., nuclear, chemical, fuel

facilities), place communities at risk; • Analyze environmental factors quickly and

effectively;

• Merge real-time sensor output (video, biochemical sensors) with archived geospatial data;

• Identify, characterize, and analyze a wide

variety of risks to transportation networks

through a gradual program of gathering image

intelligence along rights-of-way;

• Create detailed maps of an area that has

suffered attack to assist in response.

The September 11 attacks have forced a reevaluation, not only of the security aspects of

U.S. critical transportation infrastructure, but also

of the nation's institutional policies toward access

to and use of information. For example, current

policies governing the collection, use, and

sharing of critical geospatial information inhibit

the most efficient use of these data in developing

mitigation procedures and in responding to

terrorist threats and actions. On the one hand, the

potential for misuse of critical information has

caused federal and state agencies to restrict access to information by removing it from the

Internet and other public venues. On the other,

combating the threat of terrorism and responding

to future attacks will require more effective

sharing and use of geospatial data and

information. It will also necessitate within the

geospatial community continued improvement of

analytical geospatial methods and software

directed at improving transportation security.

Accordingly, the nation needs new institutional

policies to support coordinated efforts in support

of improved transportation security and

coordinated emergency response. The need for

interoperability of communications systems is widely understood;

9 the need for interoperability

of geospatial information is less well understood.

Workshop participants recognized that

meaningful progress toward preparing the nation

to prevent and respond to attacks on elements of

the nation's transportation networks required the

coordinated effort of agencies across the federal

government, among federal, state and local

governments, as well as among government and

private sector geospatial data providers and

analysts. For example, it is little recognized by

the general public that in addition to destroying

two large buildings and killing nearly 3000

people, the attacks on the World Trade Center

severely affected New York's transportation

lifelines, making rapid response by emergency

workers especially difficult. Throughout the

cleanup process, transportation routings changed

daily and even hourly, making timely map

updates necessary. Although the agencies and

individuals who responded to the challenges of

saving lives and cleaning up after the attacks

accomplished wonders despite having no

advanced warning and working in extremely

dangerous conditions, the lack of established

coordination plans slowed their efforts. In effect

the teams "made do" with what they could cobble

together in terms of base data and imagery

acquired by daily aircraft over flights, which

commenced on September 14, three full days

after the attacks.

10 Fortunately, thanks to the

many individuals and organizations willing to help, they were eventually able to provide

coordinated mapping, GPS, and imagery

information in a rapid fashion.

11 This

coordination was more a reflection of individuals

committed to dealing with a massive problem,

Q Statement of Bruce Baughman Office of National

Preparedness, Federal Emergency Management Agency,

before the Committee on Transportation and

Infrastructure, Subcommittee on Economic Development,

Public Buildings, and Emergency Management, U.S. House of Representatives, April 1 1, 2002.

10 Jerry Dobson, "How Will GIS Help Fight the War on

Terrorism?" GEO World: liltp: //www. senplace.com,

November2001

.

11 Bruce Cahan and Matt Ball, "GIS Ground Zero: Spatial

Technology Bolsters World Trade Center Response and

Recovery," GEOWorld: http://www.scoplace.com,

January 2002.

302
than the result of prior planning by the

responsible local, state, or federal agencies. These

organizations could have put teams in place and

worked more quickly and effectively if

institutional policies of their various agencies had

encouraged the sharing of information and

responsibility for response.

The geospatial data and information developed

for other uses can also support improvements in

transportation security. One of the strengths of a

geospatial information systems approach is that one comprehensive database for a region can

support many different applications. For example,

as several workshop participants noted, much of

the information developed by state and local

communities to support infrastructure planning can be used to mitigate the threat of terrorist

attacks or to enhance the speed and effectiveness

of emergency response teams in case of attack or

natural disaster. However, to be fully useful in

combating potential terrorist threats, the structure

and content of geospatial databases, including

remotely sensed imagery, will have to be

reviewed for their applicability in improving

transportation security. In some cases, the

existing framework and transportation databases

will need to be enhanced.

The need to improve critical infrastructure

protection has created new requirements for

remote sensing and GIS analysis. In order to reap

the benefits of data and information from nontransportation sources, it will be necessary to develop analytical software designed especially

for the security mission. Such software and

methods need to be widely available to provide

comparable analytical capabilities across the

United States. New analytical methods in RS/GIS

specifically devoted to the support of critical

infrastructure protection could greatly strengthen

the nation's ability to improve transportation

security. Particular attention should be given to a

significant, but limited number of critical

transportation assets (e.g., pipelines, bridges,

ports, inter-modal facilities), rather than

attempting to develop functionalities that

encompass everything.

Some remote sensing methods that have been

developed for military applications may be

transferable to civilian security use. However,

civilian measures must consider economic, social,

and legal aspects that are be less relevant in a purely military situation. The measures must be

designed for high reliability. Needs include: • New methodologies and research to identify, assess and monitor high value, highly vulnerable

transportation facilities across the United States;

• New methods of analysis and use of

sensor and imagery assets to assess

vulnerabilities, monitor assets, and aid in postevent recovery. For example, sensitive non- imaging sensors that can detect chemical,

explosive, or radioactive plumes could help in

monitoring container cargo ships from aircraft or bridges under which all ships in a port must pass.

Remote sensing technologies could be a major

asset in identifying and mitigating transportation

security weaknesses throughout the United

States. In the past, security has not been a primary consideration in transportation planning.

The U.S. transportation system as it now exists

possesses many vulnerabilities that could have

been avoided with careful advanced planning. For

example, in some cases, critical infrastructure

elements have by chance been co-located with

high-value, vulnerable facilities, such as petroleum refineries or power plants. Such colocation has been a matter of convenience or of a previous lack of concern about terrorist attack.

Remote sensing technologies can in time assist in

reducing some of these vulnerabilities. Over the

longer term, the need for instituting extensive

security measures can be reduced by preemptive

planning of lifeline facilities to minimize the

criticality of individual facilities to the network.

The planning process should emphasize

decentralization of facilities and redundancy of

the functions of those facilities. In this way

individual facilities, be they pipeline corridors or roadway conurbations, are reduced in their

critical role in the network and their

attractiveness as targets. Future planning and

implementation of transportation lifeline facilities

must have security as a primary consideration.

The United States needs to develop an accessible

geospatial infrastructure corresponding to, and

compatible with, the nation's transportation

infrastructure. The resulting geospatial

303
information should reflect all elements of the

transportation infrastructure, and include detailed

information on location, structure, condition, and

other attributes. This information should be

broadly accessible to transportation and security

professionals. Improving the interoperability of

transportation geospatial databases should be a high level priority in the attempt to enhance U.S.

transportation security. Currently, in attempting

to use transportation databases, users often

experience limitations on availability, integration,

and use of geospatial data and technologies for

transportation security. Workshop participants

pointed out that the information regarding the

nation's information infrastructure is widely

dispersed in a variety of databases, in a

multiplicity of formats and software. Hence,

many of these databases are not readily

interoperable, making the task of using them

especially difficult in times of crisis.

In particular, there is a lack of suitably

interoperable technical standards, both for data

sharing and for the operation of hardware and

software. Although the Federal Geographic Data

Committee (FGDC) has established federal

standards for preparing and sharing geographic

data, which have done much to improve

interoperability of geospatial databases on the

federal level, much more needs to be done on the

state and local level to reap the benefits in times

of crisis. State and local databases often lack

sufficient interoperability in emergency

situations, whether caused by natural disasters,

human error, or deliberately by terrorists. Yet,

most emergency situations involving

transportation systems, either actively or

passively, will involve state and local agencies

and jurisdictions. The FGDC has recommended

several actions that, if accomplished, would

sharply improve the nation's security.

12 If

extended to transportation, these

recommendations would under gird the U.S.

transportation security.

4.3 Value of Remote Sensing for Critical

Infrastructure Protection

"Homeland Security and Geographic Information

Systems,"

http://vvwvv.fgdc.gov/publications/homeland.htTnl .

4.3.1 Imaging Sensors

Scalability: Literally hundreds of civil

applications make use of remotely sensed

imagery in a variety of analytical products. A

multiplicity of image types has evolved to address specific applications, and many can also

be used for transportation lifeline protection, if

ingested and integrated into shareable databases.

Imagery can be acquired across many scales and

corrected geospatially to fit agency-specific map

projections. For incident management, planners

already have tools available to them for

integrating coarse with fine resolution satellite

views, and with more detailed high, medium, and

low altitude aerial data sets from LiDAR and

RADAR systems. The large number of

demonstrated applications meeting industry and

government standards should be systematically

reviewed for their use in transportation lifeline

security.

Spatial Relationships, Vulnerability, and UpDating: Urban planners have traditionally

recognized the importance of tone, texture, size,

shape, and spatial arrangement in analyzing

landscape features. For transportation lifelines,

most applications focused on right-of-way

planning, engineering cut and fill measurements,

feature identification, and land use classifications

for map-making purposes. The imagery was

generally archived or even destroyed after its immediate use was achieved. Where these

archives can be retrieved, they represent a resource for measuring the rates and directions of

lifeline infrastructure growth. These images also

show the changing spatial relationships among

urban features that might not be considered by

planners until the post-recovery phase of an

incident (for example, obstructions like fences

between buildings and roads that would prevent

rapid evacuations, or road designs that have

changed from two-lanes to four-lanes with a median). Most important, time series imagery is scalable from the local level for which it was

acquired, to smaller scales for which it will be

required for incident management. In short,

lifeline vulnerabilities develop over time and can

only be efficiently catalogued, referenced, and

304
assessed by examining the history of imagery of

an area.

Data Mining & Visualization: For the more

sophisticated needs of the 21

st Century,

transportation planners should recognize that

time-series digital images represent a data archive

that can be used to answer questions not yet

posed. They are essential for responding quickly

and efficiently to an emergency. Today's

technology allows image archives (also called

image pools) to be highly compressed for digital

storage, and for them to be queried constantly for

phenomena that are "out of the ordinary." Data

mining algorithms can burrow through a set of

chronologically arranged images (pixel-by-pixel

if need be), and can be programmed to detect

specific features, relationships, or trends that

"trigger" an event or suggest locations where

field personnel might visit. To take an example

from the environmental realm, the ability to detect the future onset of El Nino episodes is enhanced by daily, seasonal, and annual data

mining queries of sea surface temperatures in the

equatorial Pacific. Future "episodes" of El Nino

are being predicted on the basis of past occurrences melded with our growing

understanding of the triggering mechanisms.

Mathematical techniques such as rule-based

systems and fuzzy logic help the data mining

algorithms to interpret the intelligence contained

in the images. We are, in effect, "teasing"

information out of data in much the same way an

accountant can find aberrant numbers in a column

of numbers.

Visualization: Exciting opportunities abound for

visualizing aspects of transportation lifelines.

Visualization usually involves integrating digital

elevation models (derived from either stereo

photographs, LiDAR, or Synthetic Aperture

Radar) with imagery and other geospatial data.

Airport glide path obstructions, inter-modal

facilities, underpasses, overpasses, flyovers,

bridges, pipelines, international border crossings,

port facilities, and railroad crossings are all candidates for 3-D visualizations and virtual

reality, once the proper data structure and

environment have been established. For several

years, military pilots have trained for missions in

new settings by "flying over" an area virtually

using imagery gathered by satellite, draped over a

digital elevation model of the area. For lifeline

security planning, this capability alone may help

thwart or mitigate incidents involving shipments

of hazardous materials, or interdict possible

terrorist activities, by showing swat teams the

entire incident area before deploying human

resources (for example, height of roadside

embankments, locations and sizes of culverts,

view sheds, and related incident attributes.

Visualizations can, of course, be created without

image backdrops, and for some needs these may

even be desirable; but for realistic, real-time,

incident management, the actual ground area needs to be modeled.

Detecting Subtle Landscape Changes: Change

detection over time is one of the major

advantages of remote sensing technology. Two

change detection applications that have

widespread appeal for lifeline safety and proven

adaptability for incident management are detection of thermal patterns and subsidence

zones. The former represents mature technology

in use for the last 30 years or more, particularly in

fire mapping. Hot spots at WTC ground zero, were mapped in the days after September 1 1 and

were used in recovery operations to direct ground

crew operations. Figure 2b displays an image of

these spots, collected by EarthData International.

These hot spots are detectable in the 3-5/rni

spectrum because they are much hotter than their

surroundings. For detecting specific phenomena

in the ambient landscape a sensor operating in the

8-14jUm spectrum would be used (for such

phenomena as ship wakes, contrails, cool spots

left by parked vehicles), all of which represent

intelligence for counter terrorism or illegal

activities.

Detecting subsidence patterns is a less mature

technology, but one having proven transportation

lifeline safety applications that should adapt well

to incident management and planning.

Subsidence occurs frequently in areas where

ground water withdrawals exceed recharge on a seasonal or annual cycle. Subsidence is also a

characteristic of active earthquake zones. Interferograms produced from time series of

synthetic aperture radar phase data are capable of

detecting subsidence on the order of centimeters;

305
enough to endanger pipelines, rail lines, and

highways, and enough to damage bridges and

make on- and off-ramp speeds unsafe. Figure 2a

shows an area near Las Vegas, Nevada where

seasonal subsidence occurs. Monitoring Consequences of Incidents:

Recovery from damaging incidents, natural or

otherwise, is an ongoing element of security and

public safety planning. Relief and health officials

need to know where the transportation lifelines

have been disrupted, which avenues are still available for evacuation, or where evacuees are beginning to congregate. Image analysis within a GIS architecture can assist to develop this

information. In the case of a hypothetical airport

incident the use of high resolution imagery could

be used to provide overall situational awareness

to supplement ground intelligence. Figure 3

illustrates the capability of the new quick bird II

satellite system from Digital Earth that has a panchromatic resolution of 62 centimeters.

4.3.2 Non-Imaging Sensors

Detect Precise Object Locations at a Precise

Time: The constellation of global positioning

system (GPS) satellites is a well-known and

widely used capability for pinpointing the

locations of objects on Earth and for navigating

between points. A

companion capability develops data from an

inertial measurement unit (IMU) to translate

platform location to precise object locations on

the ground. In day-to-day office environments,

object locations can be adequately derived from

using imagery analyzed by soft-copy

photogrammetric techniques and integrated with

traditional GIS datasets; but incident management

may require the determination of spot locations, or to the ability to follow objects moving along

lifelines. These capabilities are possible and need

to be incorporated into the stable of operational

techniques employed at local and regional levels.

Because the GPS satellites also broadcast precise

time signals, they are also highly useful for

recording the precise time of events, which may

be critical in certain applications.

Detect Biological and Chemical Agents: Some

remote sensors are designed to detect and

ana/lyze the chemical constituents of gases

emanating from objects. Recently developed

technologies also include techniques capable of

detecting trace amounts of gases (in the parts per

million range, and finer). These are non-imaging

multi-spectral sensors that can be "tuned" to find

specific chemical compounds in a complex

atmosphere of numerous gases. Sometimes called

"sniffers," such devices could be deployed over multimodal transfer points and international

border crossings to identify suspicious containers,

railroad cars, or trucks. Harbor facilities might

also monitor movements of ships and other

watercraft by installing sniffers on uninhabited

aerial vehicles (UAVs) that can stay aloft for

anywhere from a few hours to several days or weeks. Within territorial waters, these devices

might also be mounted on helicopters for

monitoring ships well before they enter harbor

areas.

4.4 Transportation Infrastructure Treat

Reduction

Transportation organizations at all levels face

unprecedented challenges today. Increasing

demands are being placed on agencies to preserve

the existing transportation system and to take on

new missions of improved system safety and

security. A variety of advanced technologies are available to enhance planning, designing,

managing, operating, and maintaining all modes

of transportation. Aerial and satellite remote

sensing represents one area of rapid development

that can be leveraged to address these challenges.

These technologies have significant and unique

potential for application to a number of cross

cutting transportation security issues.

Advances in geospatial sensors, data analysis

methods and communication technology present

new opportunities for users to increase

productivity, reduce costs, facilitate innovation

and create virtual collaborative environments for

addressing the challenges of security

improvement for all modes of transportation.

Sensor developments include a new generation of

high-resolution commercial satellites that will

provide unique levels of accuracy in spatial,

spectral and temporal attributes. An example is shown in figure 3 using data from the Digital

Earth's commercial Quick Bird satellite to image

306
a portion of Reagan National Airport at a spatial

resolution of sixty-one centimeters. This

provides a degree of detail not available

previously to the civil transportation community

from space based sources. In addition to the high

resolution panchromatic imagery illustrated

above there are a number of other commercial

imagery products that are potentially applicable

to transportation. They include air borne and

satellite radar, LIDAR, multi-spectral, and hyperspectral sensors. There is a range of spectral and

spatial resolution capabilities for each of these sensor systems. Part of the challenge is matching

the best sensor to the specific transportation

related application.

Visualization and advanced data analysis

methods are also important capabilities.

Automated change detection within a defined

sector is one example of analysis capability that

will assist in detection of unauthorized intrusion

events. A specific application of these techniques

to transportation security is unauthorized right of

way intrusion on pipeline systems. Pipelines

often cover thousands of miles and are located in

remote areas that are difficult and expensive to monitor. Figure 4 illustrates the use of satellite

imagery and target identification analysis to detect unauthorized intrusion onto a pipeline

right-of-way in a remote area of Canada.

These geospatial technologies and data analysis

capabilities can also be applied to the evolution

of other new methods and approaches for

enhancing the safety and security of critical

transportation systems and infrastructure. The

geospatial area is a family of technologies that is in transition. The speed of geospatial product

delivery will continue to get faster while the cost

goes down. As the use of geospatial produces

increases there will be an expansion in new

software tool development. There will also be

continued improvement in imagery resolution and

the quality of derived products. There are a number of transportation related applications of

geospatial technology. These include: site

assessment and investigation, monitoring critical

transportation corridors, designing security

systems, monitoring critical infrastructure, rapid

change detection over wide areas and security

data and system integration.

There remain a number of challenges that may

slow or impede the application of geospatial

technologies to the transportation sector. These

include the need for improved methods and

authorities for better data sharing across

institutional boundaries. The developed and user communities need to communicate better and

overcome some significant disciplinary

differences. There are also some very

challenging technical issues in the multi-sensor

data fusion area to be overcome. Finally, there is a need for a focused interdisciplinary effort for

the development of geospatial transportation

priorities and the identification of specific

product requirements.

5.0 Geospatial Information Examples in New

York

When terrorists attacked U.S. targets on

September 1 1

, America witnessed one of the

worst days of its 225 year history. It also

witnessed the best of the American spirit as people everywhere rushed to help. From giving

blood to removing rubble to providing counseling

and housing - we helped one another cope. And

the geo-spatial community was no exception.

Individuals and corporations lent their spatial

expertise to those managing emergency situations

in New York and Washington D.C. The form of

this help was broad. Some staffed situation rooms. Others acquired aerial or satellite

imagery or processed and interpreted data. For

some, helping involved merging data from

various agencies into coherent situational maps.

Among the first spatial data providers into the

breach were those providing satellite-based

imagery.

Only three hours after the two jets crashed into

the Twin Towers of New York's World Trade

Center, SPOT Image's SPOT 4 satellite was over

the East Coast. It used its middle-infrared

capability to capture 20-meter resolution infrared

images of the fires blazing in Manhattan. SPOT

provided these images to government agencies

and posted on its Web site free of charge. A

government client then tasked SPOT with

capturing imagery every day following the event.

307
SPOT accomplished this task using a combination of its satellites. It also posted

samplings of these images on its Web site and

distributed them free of charge. On September

12, Space Imaging's Ikonos satellite acquired 1- meter imagery of both the WTC area and the

Pentagon (see Figure 5). That company donated

its imagery to many state and federal agencies,

including FEMA, U.S. Department of

Transportation, the New York Governor's office,

the New York/New Jersey Port Authority, and

more. Also, like SPOT, Space Imaging offered

imagery at no charge on the Internet, hoping to help the world understand the devastation. In just

eight days, more than 720,000 of the 2 MB

images were downloaded and more than 1 Terra

Bytes of data was served. The imagery provided

by these two firms were the only overhead views

of the disaster area available for several days, as airplanes and helicopters were not allowed to fly.

On September 14, the New York State OFT in

Albany asked EarthData to assist in recovery

efforts at the WTC. The mission, jointly

developed by OFT and EarthData, was to conduct

a series of overflights above what had by then

been dubbed Ground Zero. A combined suite of

three sensors collected black-and-white digital

imagery at 6-inch resolution, elevation data with

a vertical accuracy of 6 inches, and thermal

information that can detect changes in surface

temperature of less than 1 degree centigrade.

13

The combined geospatial information collected

by these sensors was converted into a range of

computer maps and application products that

recovery effort personnel used to accurately

define the position of underground gas and

electric lines that were in close proximity to the

disaster area, calculate movement and cubic

volume of all rubble piles, monitor the

development or retardation of the underground

fires that burned for weeks, help develop

structural audits of all damaged buildings in the

vicinity, and develop debris removal plans. These

capabilities and products helped the recovery

plan move forward efficiently while ensuring the

safety of personnel on the site. The data garnered

13 FEMA Press Release, WTC Disaster Response Status Report

FEMA-1391-DR-NY-PR-107, March 4, 2002

by EarthData was rushed daily to command

centers in Albany and lower Manhattan and

distributed immediately to firefighters, engineers,

city officials, and other rescue workers. Among

the data collected were digital aerial imagery at 6-

inch resolution acquired from 5,000 feet AGL

(see Figure 6). Figure 7 is digital aerial imagery

of similar resolution collected before the disaster

occurred. EarthData converted the images to map form (horizontally accurate to 53 feet)

through digital orthophotography. The resulting

image showed all ground objects in their correct

geographic position, enabling planners and rescue workers to obtain precise measurements of

distances between objects as well as accurate

computations of area. By showing current

conditions at the site, these maps provided

planners with tools to direct equipment and

personnel.

Also flying at 5,000 feet AGL, EarthData

acquired LIDAR (light detection and ranging

system) elevation data for the WTC beginning on September 15 (see Figure 8). It acquired data

daily for several weeks, then continued with

weekly mapping thereafter. Among the products

flowing from the LIDAR data were topographic

maps created by a team at the Center for the

Analysis and Research of Spatial Information at Hunter College. These topographic products

revealed craters that dropped 30 feet below street level and rubble that rose more than 60 feet

above. Rescue personnel consulted the maps to identify areas prone to shift or collapse when

navigating the mounds of debris. A civil

engineering unit from the New Jersey Institute of

Technology (NJIT) helped position planes

hovering above the site. Equipped with advanced

LEDAR technology, working in concert with

GPS, the planes provided data from which threedimensional digital aerial images and maps were

produced. Recovery crews on the ground used the

maps to assess damage and assist in orchestrating

the clean up. To monitor the fires that burned for

weeks within the rubble, EarthData used a thermal sensor flown from 3,000 feet AGL.

Figure 2b is a computer composite of an

orthophoto map image (horizontally accurate to 53 feet) of the WTC site acquired on September

17, 2001 combined with a thermal image.

EarthData generated the color composite overlay

308
using a thermal sensor that is sensitive to infrared

radiation rather than light.

14 Thus, it revealed the

location of hot spots within the debris field,

indicating a strong probability of lingering

underground fires.

As with so many other activities in the wake of

the attacks, the imaging mission was enabled by

selfless teamwork and a volunteer spirit. ERDAS,

Inc, for instance, donated software to produce

maps. Hurley and Associates, Inc. of Maryland

donated the thermal imaging system. The

National Oceanic and Atmospheric

Administration's National Geodetic Survey

contributed staff and equipment to operate the

GPS base station that was used to reference the

aerial imagery. And the New York State OFT

coordinated flight clearances and donated office

space for personnel to produce maps based on the

aerial information. But imagery was just one of

many tools used to manage activities following

the disaster. As with other geospatial challenges,

imagery was only part of the information needed

to respond to this disaster. It was used in

conjunction with many other technologies and

information sources, including, ground-based

intelligence and GIS. Getting transportation

moving again was also critical to response

efforts. With subway tunnels destroyed, roads

closed and normal transit routes disrupted, New

York City was able to integrate data from several

city departments into a single map (see figure 9).

With these tools they were able to help people

make better-informed decisions on how to restore

and rebuild transportation routes. GIS provided

the ability to create maps that helped rescue workers know where to look for subway

entrances and water mains under all the rubble.

The New York State Technology Office used GIS

to provide other state officials with information

on where every state office was located in

Manhattan. State officials also deployed

databases to track volunteers and to provide

information to blood donors and those seeking to donate money or supplies.

The New York response also illustrates how GIS

can play a crucial role in infrastructure

management of an emergency. Following the

terrorist attacks, GIS companies were on site at or near ground zero to assist in damage assessment,

development of evacuation plans and mapping of

critical resources. One example of a 3D GIS

product to characterize the site and allow for

analysis and fly-through assessment is shown in

figure 10. GIS helped get respohders to specific

sites and aided in identifying and locating areas of interest for rescue and recovery teams. An

Emergency Command Center was set up at Pier

92 and personnel from various companies were

consolidated at that location to handle requests

from the field and provided rapid response data.

6.0 CONCLUSIONS

It is clear that despite excellent efforts by many

groups the approach to providing information for

infrastructure management in emergencies is not

often effectively utilizing a wealth of data that

resides, with various organizations. This existing

information and technology could provide

mangers important information products that

could save lives, reduce damage to property, and

lessen environmental impacts. The current

situation is characterized by numerous

shortcomings that inhibit optimal decision

making for emergency management. The

inability to access information and the lack of

standardization, coordination, and

communication are all obstacles that need to be overcome. In addition there are tools in the

geospatial information area that could contribute

immensely to infrastructure management in

emergencies and in more routine operational

situations. One of the ways to accomplish these

aims is to make more effective and timely use of

aerial and satellite surveillance and monitoring

technologies. The challenge is how to adapt the

— Page, Daniel, "GIS played a critical role in the aftermath of

14 Harwood, Susan, "Creating a Disaster Management September 's attacks ", Mobile Government, Washington, DC,

GIS on the Fly ", ESRI New York City Office, January January, 2002

2002

309
existing capabilities of image-derived intelligence

in the short-term to fit new operating

circumstances, and what new ways of doing

business will we need in the long-term to develop

an infrastructure that is resilient to various forms

of attack.

Advances in Information systems, satellite

imaging systems and improved software

technologies have led to opportunities for a new

level of information products from remote sensed

wide situational awareness over a broad area could also be addressed using remote sensing

data sources. In the area of spectral sensing new

hyper-spectral sensing systems will allow

detection and identification of specific surface

materials, air and surface pollutants and ground

vegetation types. All of these information

products can be useful in the response, recovery,

and rehabilitation phases of infrastructure

management preparedness.

The strength of any paradigm shift is realized if

new questions can be answered along with

traditional ones. Using remote sensing

technologies combined with information sharing

will address many, if not most, traditional safety,

hazards, and disaster issues facing infrastructure

managers, as well as provide for those raised by

data. The integration of these new products into

existing disaster response systems can provide a wide range of analysis tools and information

products that were not possible before. For

example, with the higher resolution imagery

change detection for building damage assessment can be conducted rapidly and accurately with

these new data sources. Transportation

infrastructure and system

deliberate acts of aggression. Despite

technological advance, disaster risk continues to grow. Infrastructure emergency managers and

others continue to be called on to make decisions

during disaster events, as well as in the pre-andpost disaster phases, with incomplete

information. In order to make optimal decisions

to reduce the loss of life and property,

stakeholders uniformly must be able to obtain the

needed information in a format that is appropriate

for their capabilities. There is also the need for

parties at great distances from each other to be

able to share information in a seamless fashion

that also allows the shared information to be

interactive with local data and be used to create

new integrated products tailored to the situation.

310
Estimated Extent of America's Transportation System

(Source: DOT/RSPA)

System ** Infrastructure Use

210 million vehicles 4 million miles of road

000 bridges (1/3 of 127,000 urban

>ridges serve as major arterials

4 trillion passenger-miles

920 billion freight ton-miles

XX railroad cars 150 intercity Amtrak trains 150,000 track miles20 million passenger miles

1.3 trillion ton-miles

175,000 aircraft 5,500 public use airports 403 million passenger miles

12 million watercraft 25,000 miles of waterways 600 billion ton-miles of oil Pipelines 1.6 million miles 19 trillion cu.ft. of gas

:OMEX/TEU (Multimodal) 7reighters, trucks, railroad cars 5 million each year through major U.S. ports

Table 1: Summary of America's Transportation System

311
Figure 1 . Sample LiDAR data set of the downtown New York area collected by EarthData.

The vertical accuracy is about .8 feet on average across the data set.

Figure 2a,b. Rainbow pattern at left shows land subsidence of a few centimeters near Las

Vegas, NV. Such patterns can disrupt rail, pipelines, and road networks, and cause overpasses

and bridges to be unsafe. The orange colors at right represent hot spots at the WTC "ground

zero 5 weeks after the attack. Such patterns assist post-incident recovery.

312
313
Figure 5: Space Imaging IKONOS pan-sharpened one Figure 6: Six-inch digital photograph near the disaster area meter spatial resolution image ofthe World Trade

Center taken on September 15, 2001

Figure 7: Digital photograph prior to disaster Figure 8: LIDAR data showing debris location

Figure 9: GIS map product of transportation routes Figure 10: 3D GIS product showing color-coded buildings

314
Risk Assessment on Storm Surge Flood

by

Kenichi Torii

1

, Fuminori Kato

2

ABSTRACT

The method of risk assessment for hazard map

on storm surge flood has not been established

completely. For nationwide distribution of the

hazard map, several research subjects in process

of the risk assessment on storm surge flood will

be investigated in this study. Probabilistic

evaluation on tide level and wave overtopping

rate, judgment on coastal dike break, and flood

simulation are key elements in the risk

assessment.

KEY WORDS: Flood, Risk Assessment, Storm

Surges

1. INTRODUCTION

Typhoons have caused catastrophic damage by

flooding due to storm surges in Japanese history.

As shown in , large scale of storm surge disasters

have occurred frequently until 1961. For example,

about five thousands of people was dead with

storm surge by a typhoon in 1959, which was

called Ise Bay Typhoon. After 1970's, catastrophic

flood disaster due to storm surges have been rare.

There are two main factors of small number of

severe storm surge flood in recent 40 years. One

is construction of coastal dike against storm surge,

the other is improvement of weather forecast

system. These factors made residents in coastal

area feel as if their houses had been safe from

storm surge flood.

Contemporary structure of great cities in coastal

area is a complex including ground space and

underground space. Storm surge larger than

designed level in shore protection works may

cause severe flood disaster of unexpected process

in the great cities.

One of effective countermeasures against storm

surge disaster is hazard maps on storm surge,

however, the maps have not been distributed

overall in Japan. The method of risk assessment

for the map has not been established completely.

This study is to establish the method of risk

assessment on storm surge flood.

2. FRAMEWORK OF RISK ASSESSMENT ON

STORM SURGE FLOOD

There are some matters for making hazard maps

on storm surge flood as shown in Figure 1

.

One is how to set tide level and wave overtopping

rate as boundary conditions of flood simulation.

Return period of high tide level has not been

evaluated accurately because of short period of

tidal observation. Dependency of wave height on

1 Head, Coast Division, River Research Department, National Institute for Land and Infrastructure

Management, Ministry of Land, Infrastrucuture and Transport, Asahi 1 , Tsukuba 305-0804, Japan

2 Senior Researcher, ditto

315
tide level also makes it difficult to evaluate return

period of wave overtopping rate.

Second one is how to judge coastal dike break.

Most of coastal area in Japan is protected from

storm surges and high waves by coastal dikes in

designed level. Although most of coastal dike is armored with concrete or asphalt, tide level and

wave higher than designed level may break

coastal dike.

From the viewpoint of mitigating fatalities, risk

assessment on storm surge flood can be

conducted as shown in Figure 2.

(1) Probabilistic evaluation on tide level and wave

overtopping rate

Tide levels and wave overtopping rates are

necessary for flood simulation in coastal zones as boundary conditions. Return period of tide levels

and wave overtopping rates can be evaluated on

results of numerical simulation with a

probabilistic generation model for parametric

properties of typhoon.

(2) Judgment on coastal dike break

Break of coastal dike increases inflow rate of

seawater awfully. The break can be judged based

on wave overtopping rate and type of coastal

dike.

(3) Flood simulation

Inundation depth, velocity of flood flow, and

arrival time of flood in coastal zone can be

estimated with a numerical model.

(4) Risk assessment

From the viewpoint of need and safety of

evacuation and safety of houses, risk of storm

surge flood can be assessed based on results of

the flood simulation.

3. PROBABILISTIC EVALUATION ON TIDE

LEVEL AND WAVE OVERTOPPING RATE

Probabilistic evaluation on tide level and wave

overtopping rate can be conducted as shown in

Figure 3. First is making a probabilistic

generation model for parametric properties of

typhoon based on recorded data. Second is evaluating return period of tide level and wave

overtopping rate based on Monte Carlo method.

Probabilistic generation model for parametric

properties of typhoon can be made in methods by

Yamaguchi et al.(1994) and Hashimoto et al.

(2001). In their method, probability of number,

location and its variation, atmospheric pressure

and its variation, and radius of typhoon are evaluated based on data of typhoon for the past

fifty years.

In evaluating return period of tide level and wave

overtopping rate, typhoons are supposed with the

model for hundreds of years. After that, tide level

and wave overtopping rate due to each supposed

typhoon are calculated through numerical

simulation. Finally, return period of tide level and

wave overtopping rate are calculated.

Figures 4 to 9 show results of a typhoon

simulation with the model. The simulation for

past fifty years was conducted fifty times. It corresponds a simulation for 2,500 years.

Although calculated standard deviations of

atmospheric pressure at typhoon eye, typhoon

speed and typhoon direction were smaller than

observed ones, calculated average values of them

were almost equivalent to observed ones.

4. PROBABILISTIC EVALUATION ON

COASTAL DIKE BREAK

Although most of coastal dike is covered with

concrete or asphalt in Japan, coastal dike might be

broken due to excess waves and/or cavern in fill.

316
Table 2 shows an example of wave overtopping

rate critical to destruction of dikes and

revetments.

Probability of coastal dike break should be

evaluated for each wave overtopping rate based

on breach examples and results of physical model

experiments.

5. RISK ASSESSMENT BASED ON FLOOD

SIMULATION

If houses could be inundated due to a storm surge,

the residents should evacuate to designated safe

locations prior to any danger. Whether evacuation

is required can be evaluated based on the

maximum flood depth obtained from flood

analysis using the worst scenario.

On the other hand, safety for evacuation depends

on time for flood to reach each point as well as

the maximum flood depth. In a low plain field,

people in evacuation is safer at points far from the

shoreline because it takes a long time for flood to

reach there, although significant variation in the

maximum flood depth is not observed throughout

inundated area.

According to the criteria shown in Table 3, we

ranked the relative hazards to human based on the

maximum flood depth and the time for flood to

reach obtained from flood analysis. The points

where the maximum flood depth corresponds to

the flood above floor level were assumed to be

the evacuation-designated areas. Note that some

of the houses that were not inundated cqmpletely

or were not swept away even if flooded, might be

ranked D indicating that evacuation is not

required.

For the houses in the evacuation-designated areas, assuming that the residents began to escape when

seawater overflowed the embankment, a comparison was made between the time required

for evacuation and the time for flood to reach. If

the time for flood to reach was longer, the houses

were ranked C. Otherwise, the residents could

escape when the flood depth was low by using the

evacuation road, even if they began to escape at

the onset of seawater invasion.

Further hazard ranking was made based on

whether the residents could safely reach the

walkable evacuation road even if they began to escape in case of a flood. The rank C houses were

divided into subgroups based on whether

residents could reach the safe evacuation road if

they began to escape upon becoming aware of the

flood on their houses.

This method allows us to evaluate the effects of

improving evacuation locations and making the

houses water-resistant inside the embankment.

For example, if safe evacuation locations that are

not inundated even during a flood, are improved

into rank A or B locations, then the time required

to escape is reduced and the houses in the vicinity

of the locations are ranked C+ or C-.

Principal cities in Japan are in risk of riverine

flood and flood due to drainage failure as well as storm surge flood. Flood risk should be assessed

comprehensively for residents convenience.

We are now making a probabilistic generation

model for parametric properties of typhoon.

Evaluation on tide level and wave overtopping

rate will be conducted for four coasts facing

Tokyo Bay, Ise Bay, Osaka Bay and Tosa Bay,

where severe storm surge has occurred.

REFERENCES:

Goda, Y.: Study on Wave Overtopping Rates at

Seawalls, Report of Port and Harbor Research

Institute, No.9, Vol.4, pp.3-42, 1970. (in

Japanese)

317
Hashimoto, N., H. Satoh, K. Matsuura and M.

Ichikawa: Modeling of Probable Typhoon and Its

Statistic Properties, Proceedings of Coastal

Engineering, Japan Society of Civil Engineers,

pp.456-460, 2001. (in Japanese)

Torii, K. and F. Kato: Study on Storm Surge

Disaster, Technical Memorandum of Public

Works Research Institute, No.3803, 97p., 2001.

(in Japanese)

Yamaguchi, M., Y. Hatada and Y Nakamura:

Probabilistic Typhoon Model and Its Application

to Extreme Value Estimation of Storm Surges and

Waves, Proceedings of Annual Conference of

Japan Society for Natural Disaster Science,

pp.135-138, 1994. (in Japanese)

318
Table 1 Storm surge disasters in Japan

Human casua ties Damage to houses

Date Ma'or c,ama9ed

area Dead Injured Missing

(""nrnrilptplw

ucoii uycu

r Clf Lldll y ucoli uycu a vvciy

1 Oct. 191 7 Tokyo Bay 1,127 2,022 197 34,459 21 ,274 2,442

1 3 Sep. 1 934 Ariake Sea

21 Sep. 1934 Osaka Bay

373

2,702

181

14,994

66

334

1.4: 38,771

20 49,275791

4,277 Muroto

27 Aug. 1942 SuoSea 891 1,438 267 33,283 66,486 2,605

1 7 Sep . 1 945 Southern Kyushu 2.076 2,329 1,046 58,432 55.006 2,546 Makurazaki

3 Sep. 1950 Osaka Bay 393 26,062 141 17,062 1 01 ,792 2,069 Jane

1 4 Oct . 1 951 Southern Kyushu 572 2,644 371 21 ,527 47,948 1,178 Ruth

7 Sep. 1959 IseBay 4.697 38,921 401 38.921 113,052 4,703 IseBay

16 Sep. 1961 Osaka Bay

21 Aug. 1970 TosaBay

185

12

3,897

352

15

1

13,292

811

40,954

3,628

536 Second Muroto

40 No .10 30 Aug. 1985 Ariake Sea 3 16 0 0 589 0 No.13

S7" Inundation depth

wave

' overtopping

Figure 1 Schematic view of storm surge flood

Probabilistic evaluation on tide

level and wave overtopping rate

Judgment on coastal dike break

Flood simulation

Risk Assessment

Making a probabilistic generation model

for parametric properties of typhoon

Calculating tide level and wave overtopping

rate for hundreds of years

Typhoon supposed with the model

r

Simulation for tide level and wave

overtopping rate

Calculating the return period

Figure 2 Flowchart of risk assessment on storm

surge flood

Figure 3 Flowchart of probabilistic evaluation

on tide level and wave overtopping rate

319
Figure 4 Average atmospheric pressure at typhoon eye

Figure 5 Standard deviation of atmospheric

pressure at typhoon eye

Figure 6 Average typhoon speed Figure 7 Standard deviation of typhoon speed

320
Figure 8 Average typhoon direction Figure 9 Standard deviation of typhoon

direction

Table 2 Allowable wave overtopping rates

(Goda, 1970)

type crest of dike landward

slope of dike wave overtoppingrate (mA3/m/s)

1 not armored not armored <0.005

2 armored not armored 0.02

3 armored armored 0.05

321
Table 3 Hazard ranking criteria for storm surge flood (Torii and Kato, 2001)

Hazard

rank

Expected

maximum

flood depth

Houses

Expected

time for flood to reach

Safe evacuation if residents begin to escape on becoming aware of the flood on

their houses

Expected condition Desirable

action

A

above floor level

shorter than

Tl and Tr impossible

evacuation is possible if residents begin to escape before

the onset of seawater Invasion

escape

prior to the onset of seawater

invasion

B

above floor level

shorter than

Tl, but longer than TrImpossible

evacuation is possible in flood If residents begin to escape at the onset of seawater invasion

C+

above floor

level

longer than

Tl and Tr impossible

evacuation is possible In dry

condition if residents begin to escape at the onset of seawater

Invasion, but Impossible if residents begin to escape upon

becoming aware of the flood on

their houses

C- above floor

level

longer than

Tl and Tr

possible

evacuation is possible even if residents begin to escape upon

becoming aware of the flood on

their houses

escape on

self- judgment

D

above floor level

high-rising, water-resist safe in their houses

stand by in

their houses

E below floor level safe in their houses

(none) no

inundation safe In their houses

Tl: Time for residents to reach an evacuation location Tr Time for residents to reach a walkable evacuation road

322
MANUSCRIPTS AUTHORED

FOR PANEL MEETING

BUT NOT PRESENTED ORALLY

Shaking Table Tests on Seismic Response Reduction Effects

of Rocking Building Structural Systems

by

Tatsuya Azuhata1

, Mitsumasa Midorikawa2, Tadashi Ishihara

3 and Akira Wada4

ABSTRACT

Shaking table tests are carried out to examine

seismic responses of rocking structural systems

with base plate yielding (base plate yielding

systems) comparing with those of fixed-base

systems and simple rocking systems. Furthermore

a simple method using the equivalent one mass

system to predict the base shear that base plate

yielding systems suffer when they begin up-lifting

response (critical base shear of up-lifting

response) is proposed. It is concluded that the

base plate yielding systems can reduce effectively

the seismic response of structures and their

critical base shear of up-lifting response can be

predicted by the proposed method appropriately.

KEYWORDS: Shaking table test Rocking system

Base plate

Earthquake response reduction

Smart structure

1. INTRODUCTION

It is pointed out that effects of rocking vibration

(up-lifting response) can reduce the seismic

damage of buildings subjected to strong

earthquake ground motions [1][2]. Based on this

knowledge, we are now developing the rocking

systems that can cause rocking vibration under

appropriate control during earthquakes [3].

The rocking system can be regarded as one of

smart structural systems. Smart systems are defined as structural systems with a certain-level

of autonomy relying on the embedded functions

of sensors, actuators and processors that can

automatically adjust structural characteristics, in

response to the change in external disturbances

and environments, toward structural safety and

serviceability as well as the elongation of

structural service life [4]. Although the rocking

system has neither specific devices nor computer

control systems, it satisfies this definition. It is thought the rocking system is one of the simplest

smart structural systems.

One of the rocking systems we are developing has

weak base plates at the bottom of each steel column

of the first story. When the weak base plates yield

by tension force of the column during a strong

earthquake, the building causes rocking vibration.

In this paper, shaking table tests are carried out to examine seismic responses of this type of rocking

systems (the base plate yielding systems) comparing

with those of the fixed-base systems and the simple

rocking systems which are fixed only in the

horizontal direction under the footing beams.

1 Senior Research Engineer, Building Research Institute, Tachihara 1, Tsukuba-shi, Ibaraki-ken, 305-0802,

Japan

2 Research director for International Codes and Standards, BRI

3 Section Chief, Building Guidance Division, Housing Bureau, MLIT

4 Professor, Tokyo Institute of Technology

325
Outlines of the base plate yielding system and the

simple rocking system are illustrated in Fig.l.

When we apply the base plate yielding system to

real structures, we must adjust the physical

characteristics of the base plate such as those

dimensions and yield point, so that the building

structures cause rocking vibration before they

yield. Therefore we need to grasp the relation

between the physical characteristics of the base

plate and the critical base shear of up-lifting

response. In the following, a simple method to

grasp this relation is proposed and the

applicability of this method is verified using test

results.

2. SPECIMENS AND TEST METHOD

Specimens for the tests are composed of a steel

frame shown in Photo. 1 and some structural

components such as weak base plates attached to

bases of the frame. Each floor height of the steel

frame is lm and total height is 5m. In the

oscillation direction, the frame span is 2m. Each

floor mass is shown in Table 1 . The cross sections

of the members are shown in Table 2.

The base plate for the base plate yielding system

is shown in Photo. 2 and Fig. 2. Thickness of the

base plate is 6 mm or 9 mm. Material

characteristics of steels used for base plates are shown in Table 3. In the following, the specimen

with base plates whose thickness is 6 mm is

referred to as BP6 model and the specimen with

base plates whose thickness is 9mm is referred to

as BP9 model respectively.

The structural component for the simple rocking

system is shown in Photo. 3 and Fig. 3. In this

component, shock absorbers can be installed

between the upper and the lower part of it to weaken shock force which occurs when the

structure lands after up-lifting.

However, it was cleared that the test results for

this specimen were not affected by the existence

of the absorbers and the kind of them after the

tests. Thus only the results that we get when using

rubbers as the shock absorbers are shown in this

paper. The specimen for the simple rocking

system is referred to as R model.

The specimens are oscillated only in one direction,

that coincides with the strong axis direction of

columns. And in this direction, the frame shown

in Photo. 1 has no brace. The input ground motion

used for the tests is 1995 JMA Kobe NS, of which

the time scale is shorten to 1/V~3.

3. TEST RESULTS

Fig. 3 shows the relation between maximum input

acceleration and response values of each model.

Up-lifting responses are small when the

maximum input acceleration is nearly equal to

1 .5m/sec2. These values of BP6 and BP9 models

are almost 0.0. As the input acceleration becomes

larger, the up-lifting responses increase

monotonously. When the input acceleration

becomes about 3.5 m/sec2, the up-lifting

displacement of R model is 6.2 mm and that of

BP6 model is 2.4 mm.

Base shears of all models are almost same each

other when the input acceleration is smaller than

1.5 m/sec2. When the input acceleration becomes

larger than 2.0 m/sec2, these values of R model

are 40 kN to 50 kN and are constant

approximately. And base shears of BP6 and BP9

models are also smaller than those of the

fixed-base model (F model). In this study, base

shears are evaluated as the summation of each

story's horizontal force derived by multiplying

each story's mass by each story's response

326
acceleration. Each story's mass is derived by

multiplying the summation of the mass of

structural members and weights by 1.05 in order

to evaluate the mass of bolts and steel plates in

the connection.

Roof displacements of all models are almost same

when the input acceleration is smaller than 4.0

m/sec2. When the input acceleration is larger than

4.0 m/sec2, the displacement of R model is larger

than those of the other models. It is thought the

reason for this results are that the roof

displacements of BP6 and BP9 model are smaller

than those of R model and that hysteretic damping

of base plates dissipates earthquake energy even

when the specimens land.

Fig. 4 shows story shear forces of each model.

These values are derived from the same method

for the base shear. When the input acceleration is

1.5 m/sec2, the response values are almost same

each other. When the input acceleration becomes

3.5 m/sec2, seismic response reduction effects of

rocking structural systems can be observed clearly

and all story shear forces of R and BP6 models

are smaller than those of F model.

4. PREDICTION OF THE CRITICAL BASE

SHEAR OF UP-LIFTING RESPONSE

To predict the critical base shear of up-lifting

response, the moment balance of the equivalent

one mass system shown in Fig. 6 is considered.

The moment balance is expressed as Eq. (1).

Mu -a-H=M-g-(Bj^+n-Ny -B (1)

where,

Mu: the 1st effective mass in horizontal

direction

a: horizontal response acceleration which occurs when the structure begin up-lifting response

(critical acceleration of up-lifting response)

H: the 1st representative height

M: total mass

g: gravity acceleration

B: distance between columns

n: number of columns in the up-lifting side

Ny: tension yield strength of base plate

In the above equation, the first normal mode is needed to calculate the 1st effective mass Mu and

the 1st representative height H. Now, we try to

approximate this mode using horizontal story

response acceleration distribution.

Fig. 7 shows this distribution is shaped like the

reversed triangle before and after up-lifting (Input

acceleration is 15 m/sec2 and is 3.5 m/sec2). Thus

the 1st effective mass Mu and the 1st

representative height H are calculated assuming

the normal mode shape is the reversed triangle.

The 1st effective mass Mu of BP6 and BP 9

models is calculated as follows.

f

N >

2

,11, J N

i=l

= 9.83 (t) (2)

where,

m,: mass of each story

The 1st representative height H of BP6 and BP9

models is calculated as follows.

N

H = = 3-66 (m) (3)

where,

he. height of each story

The tension yield strengths of base plates of BP6

and BP9 models are calculated by Eq. (4) and Eq.

(5) respectively.

327
[BP6 model]

n-bt1 <r„ Ny =-

21 = 26.46 (kN) (4)

where,

n: number of base plates

b: width of base plate

t: thickness of base plate

/: length of base plate

ay: yield point of steel used for base plate

[BP9 model]

Ny =53.61 (kN) (5)

By substituting Mu by Eq. (2) , H by Eq. (3) and

Ny by Eq. (4) or Eq. (5) into Eq. (1), the critical

acceleration of up-lifting response a is calculated

as follows.

[BP6 model]

a = 6.21 (m/sec

2

) (6)

[BP9 model]

a = 9.23 (m/sec

2

) (7)

Furthermore the critical base shears of up-lifting

response, which are approximated as the product

between Mu and a are predicted by following

equations.

[BP6 model]

Q = Mu -a = 61.10 (kN) (8)

[BP9 model]

Q = Mwa = 90.77 (kN) (9)

In Fig. 8, the predicted values of the critical base

shears of up-lifting response are compared with

the corresponding test results. The figure shows

the predicted values rather become underestimates

to the test results, but they correspond each other

with good agreements. When the input

acceleration exceeds 5.0 m/sec2, the base shear of

B6 model tends to ascend. It is thought the reason

for these results is that tension displacements

along the longitudinal direction become

predominant under this input level. As for this

point, we need to study furthermore by

investigating the force deformation characteristics

of base plates.

5. CONCLUSIONS

The earthquake responses of the rocking system

with weak base plates (the base plate yielding

systems) were examined comparing with those of

the simple rocking system and the fixed-base

system by shaking table tests. The results of this

study are summarized as follows.

1) Base shears and story shear forces of BP6, BP9

and R models become smaller than those of F

model, in the range of input acceleration where

up-lifting response occur in these models.

2) Roof displacements of BP6 and BP9 models

are not amplified extremely even in the range of

input acceleration where those of R models

become very large. And vertical acceleration

around the frame base of BP6 and BP9 models

is smaller than that of R models.

3) The simple method using the equivalent one

mass system for prediction of the critical base

shear of up-lifting response is proposed. And its

applicability is ascertained using test results.

Acknowledgement

This work has been carried out under the

US-Japan cooperative structural research

project on Smart Structure Systems

(Chairperson of Japanese side: Prof. S. Otani,

University of Tokyo). The authors would like

to acknowledge all project members for their

useful advice and suggestions.

328
REFERENCES:

1) Hayashi, Y., Damage Reduction Effect due to Basement Uplift of Buildings, J. Struct. Constr.

Eng., ALT, No.485, 53-62, Jul., 1996 (In

Japanese)

2) Rutenberg, A., Jennings, R C. and Housner, G.

W., The Response of Veterans Hospital Building

41 in the San Fernando Earthquake,

EARTHQUAKE ENGINEERING AND

STRUCTURAL DYNAMICS, vol. 10, 359-379,

1982

3) M. Midorikawa, T. Azuhata, T. Ishihara, Y.

Matsuba, Y. Matsushima and A. Wada: Earthquake

Response Reduction of Buildings by Rocking

Structural Systems, SPIE's 9th Annual

International Symposium on Smart Structures and

Materials, 4696-33, 2002.3.

4) Otani, S., Hiraishi, H., Midorikawa, M. and

Teshigawara, M., Research and Development of

Smart Structural Systems, Proc. of 12WCEE,

2307, 2000

329
(a) Base plate yielding system (b) Simple rocking system

Fig. 1 Two types of rocking systems

Table 1 Mass of each story Table 2 Cross sections of members

Story Mass (t) RF 2.264

5F 2.292

4F 2.292

3F 2.292

2F 2.292

IF 1.456

Column and beam HI48x1 00x6x9

Footing beam H250x250x9xl4

Table 3 Steels used for base slates

Model name JIS Yield point(N/mmz

) BP6 SS400 334.18

BP9 SS400 300.86

Photo. 2 Base plate for BP model (1

)

Photo. 1 Specimen frame

Photo. 3 Structural component for R model (1)

330
> S I L*-!

-CD

"f

'

\

Fig. 2 Base plate for BP model (2)

20

E

E-15

L10

O R model -A BP6 model -H BP9 modelt

1 2 3 4 5

Input acceleration (m/sec2)

(a) Up-lifting

2 3 4 5

Input acceleration (m/aec2)

i i i

I T I

(b) Base shear

= I.I 4y

4 m til

!

i

I

i i

Fig. 3 Structural component for R model (2)

f3os

E

35

30

25

20

15

10

5

0

2 3 4 5

Input acceleration (m/sec2)

(c) Roof displacement

O R model -A—BP6 model -H—BP9 model

2 3 4 5

Input acceleration (m/sec2)

(d) Vertical acceleration around the base Fig. 4 Maximum input acceleration vs. maximum response values

331
t3

o**CO

2 -

F model O R model -A—BP6 model -0—BP9 model

10 20 30 40

Story shear force (kN)

(a) Input acceleration: 1 .5 m/sec2

50

5 -

1

-

— F model — R model -A—BP6 model |

I 20 40 60 80

Story shear force (kN)

(b) Input acceleration: 3.5 m/sec2

Fig. 5 Story shear forces

100

X F modal -A—BP6 model -S—BP9 model

2 4

Floor acceleration (m/sec2)

(a) Input acceleration: 1 .5 m/sec2

5 10

Floor acceleration (m/sec2)

15

(b) Input acceleration: 3.S m/sec2

Fig. 7 Each floor acceleration

fi=2000

^akMii'd

|( g=2000

^

Fig. 6 Equivalent one mass system

H

2 3 4

Input acceleration 0n/sec2)

Fig. 8 Prediction of critical base shears of

up-lifting response

332
Large-scale Tests for Performance Verification of Smart Structures

by

Mitsumasa Midorikawa1

, Masaomi Teshigawara', Masanori Iiba

1

, Hideo Fujitani

1

Koichi Morita2

, Tatsuya Azuhata1

, Akiyoshi Fukuda3

, Yoichi Shiozaki3, Takeshi Hiwatashi3

and Chikahiro Minowa4

ABSTRACT

The objectives of these large-scale tests are to confirm the effectiveness of several proposed

smart materials, members and structural systems.

The tests were conducted on the shaking table of

the test facility of the National Research Institute

for Earth Science and Disaster Prevention (NIED),

Tsukuba, Japan. It has a maximum displacement

of 22 cm, a maximum velocity of 75 cm/s, and a maximum weight of 500 ton. The following test

series was conducted: 1) semi-active control by

MR dampers (base-isolation system and seismic

control system), 2) damage detection system

(identification of damaged members and

evaluation of new sensors), 3) rocking energy

dissipation system (lift-up system). This paper

outlines the objectives of the large-scale tests, the

design of the test frame for each objective, and

the static and dynamic characteristics of the test

frame.

KEYWORDS: Large Scale Test, Smart

Structure, Performance Verification, Shaking

Table, MR damper, Rocking System, Sensor

Technology

1. INTRODUCTION

The Building Research institute (BRI) of Japan

and the U.S. National Science Foundation (NSF)

initiated the U.S.-Japan Cooperative Research

Program on Auto-adaptive Media (Smart

Structural Systems) in 1998 m, under the aegis of

the U.S.-Japan Panel on Wind and Seismic Effects

of the U.S.-Japan Cooperative Program in Natural

Resources. At the Joint Technical Coordinating

Committee (JTCC) meeting, research items and

plans were discussed in detail for three research

thrusts: (1) structural systems, (2) sensing and

monitoring technology, and (3) effecter

technology.

BRI planned a series of large-scale tests to

verify some smart systems developed in this

project. The effectiveness of "Semi-active

control by MR dampers", "Damage detection

system" and "Rocking energy dissipation system"

will be also confirmed.

This paper outlines the objectives of the

large-scale tests, the design of the test frame for

each objective, and static and dynamic

characteristics of the test frame.

1 Building Research Institute, 1 Tachihara, Tsukuba-shi, Ibaraki-ken, 305-0802, JAPAN

2 National Institute for Land and Infrastructure Management, Ministry of Land, Infrastructure and Transport,

1 Tachihara, Tsukuba-shi, Ibaraki-ken, 305-0802, JAPAN

3 Visiting Researcher, Building Research Institute, 1 Tachihara, Tsukuba-shi, Ibaraki-ken, 305-0802, JAPAN

4 National Research Institute for Earth Science and Disaster Prevention, 3-1 Tennodai, Tsukuba-shi,

Ibaraki-ken, 305-0006, JAPAN

333
2. OBJECTIVES OF LARGE SCALE TESTS 2.2. Damage Detection by Sensors

The basic test frame used for this large-scale test is

a 3-story steel frame. It is 3m (1 span) x 4m (2m

x 2 span) in plan and 6m high, weighs a total of 20

ton (including the weight of the 1st floor), and has

a natural period of about 0.5 sec. Some devices

for the response control system with semi-active

MR dampers and for damage detection are

installed in the middle frame in the shaking

direction. Devices for the rocking energy

dissipation system are installed in each column

foot of the 1

st story.

The NIED's shaking table (Photo 1) was used.

Figure 1 shows the performance of the shaking

table. The shaking table has a maximum

displacement of 22 cm, a maximum velocity of 75

cm/s, and a maximum weight of 500 ton. The following subjects are discussed with regard

to the large scale shaking table test.

2.1 Response Control by Magneto-Rheological

(MR) Fluid

Magneto-rheological (MR) dampers have been

expected to control the response of civil and

building structures in recent years, because of their

large force capacity and variable force

characteristics. The passive control has a

limitation of damping effect to a certain range with

frequency and input level in general. Semi-active

control reduces both response displacements and

accelerations. MR damper generates damping

force, which does not depend on the piston speed

[2]

. The target of this subject is to improve the

safety and functionality and habitability by

controlling the response displacements and

accelerations by using MR dampers.

For this purpose, MR dampers and a control

algorithm have been developed and their validity is

discussed by an analytical study and shaking table

tests.

Monitoring the structural soundness of a building

effectively reduces its life cycle cost. A series of

experimental tests for a monitoring system has

been conducted by measuring the process in which

a concrete device is broken. Devices are inserted

in the central structural plane in each frame story

(Fig. 3). The same device is installed in each

story, and the purpose of the test is to detect

damage. The table was shaken several times.

White noise excitation and micro tremor

measurement were carried out before and after

shaking.

Two identification methods were carried out

one using the data of before and after shaking and

the other using the data during shaking. For the

first, we used the flexibility method, the story

stiffness method and the identification method

using multiple natural frequency changes, etc., and

for the second we used the ARX model and

parallel processing identification method131

.

2.3 Rocking Energy Dissipation System

Rocking systems that cause rocking vibration under

appropriate control during earthquakes are now

under development141. Some researchers have

pointed out that the effects of rocking vibration can

reduce seismic damage to buildings subjected to

strong earthquake ground motions. One of these

systems has weak base plates at the bottom of each

steel first-story column. When the weak base

plates yield during a strong earthquake, rocking

vibration is caused in the building, as illustrated in

Fig. 4.

Fig. 5 shows the test frame for examining the

earthquake response reduction effects of this

rocking system with base plate yielding (base plate

yielding system). In each story of each frame,

braces made of PC steel bars are installed to adjust

the fundamental period of the test. With these

334
braces, the first natural period of the test frame

becomes 0.18 sec. In this shaking table test, 1940

El Centro NS is used with its time axis shortened by

1//"2. Thus, the first natural period of the test frame is almost equal to the predominant period of

the inputted wave motion.

3. CHARACTERISTICS OF TEST FRAME

Figure 6 shows the Fourier spectrum of response

acceleration of each story of the superstructure

excited by white noise. The roofs primary

predominant (1

st mode) frequency is 1.83Hz. This

means that the first natural frequency is 0.546 sec. However, the 2

nd spectrum value is smaller than the

3

rd value on the 3

rd floor and, the 2nd spectrum

value is larger than the 3

rd value on the 2nd floor.

This means that the secondary mode vibration is

being generated.

Tables 1-3 shows the characteristics of test

frame.

4. CONCLUSIONS

This paper has described the characteristics of a

large-scale test frame. One test frame was used for

four series of experimental tests to verify each

developing technology, such as semi-active control

for base-isolation and seismic response control,

damage detection and rocking energy dissipation.

This large-scale experimental test verified the

validity of application of those systems to smart

structures, and useful experimental results were

obtained for the settlement of future research.

Acknowlegement

This work has been carried out under the

US-Japan cooperative structural research

project on Smart Structure Systems

(Chairperson of Japanese side: Prof. S. Otani,

University of Tokyo, Chairperson of U.S. side:

Prof. M. A. Sozen, Purdue University). We

would like to acknowledge the hard work and

contribution of all members of the project for

their useful advice and suggestions.

REFERENCES:

[1] Otani S., Hiraishi H., Midorikawa M.,

Teshigawara M., Saito T. and Fujitani H.:

Development of Smart Systems for Building

Structures (Invited paper), Proc.of SPIE's 7th.

International Symposium on Smart

Structures and Materials, Paper No. 3988-01,

2000.3.

[2] Fujitani H., Sodeyama H., Hata K., Iwata N.,

Komatsu Y., Sunakoda K. and Soda S.: Dynamic Performance Evaluation of

Magneto-Rheological Damper, Proceedings

of Advances in Structural Dynamics 2000,

Vol.1, 2000.12, pp.3 19-326.

[3] Morita K, Teshigawara M., Isoda H.,

Hamamoto T. and Mita A. (2001) Damage

Detection Tests of Five-Story Steel Frame

with Simulated Damages, Proc. of SPIE's 6th

International Symposium on NDE for Health

Monitoring and Diagnostics, Paper No.

4335-18,2001.3.

[4] Midorikawa M., Azuhata T, Ishihara T,

Matsuba Y., Mtsushima Y. and Wada A.:

Earthquake response reduction of buildings by

rocking structural systems, Proc.of SPIE's 9th.

International Symposium on Smart Structures

and Materials, Paper No. 4696-33, 2002.3.

335
(a) Centralframe ofspan direction (b) Centralframe ofspan direction

for base isolation system for seismic response control

(c) Firstfloor ofbase isolation system (d) Elevation ofridge direction

for seismic response control

Fig. 2 Test frame for response control by MR dampers

336
Transducer

Crack Sensor

Controlle

Wireless

LAN

4f

SvRtem Pf!

Fig.3 Central frame of span direction for damage detection and smart sensors

in

*

Abase plale yields

3,0m

Fig.4 Outline of rocking system with base plate yielding

At*

{a) Centralframe

in span direction

(b) Edge frame

in ridge direction

(c) Edge frame

in span direction

Fig.5 Test frame for Rocking Energy Dissipation System

337
1500

Roof

4 6

Frequency (Hz) 2nd Floor

4 6

Frequency (Hz)

1500

g ^000

I s

500

0 3rd Floor A L 4 6

Frequency (Hz)

< «

100

80

60

40

20

10 0

Shaking Table

4 6

Frequency (Hz)

10

Fig. 6 Fourier spectra of acceleration response of test frame

Table 1 Masses Table 2 Characteristics of Frame Table 3 Characteristics of

(ton) Base Isolation

Roof 4.67 Stiffness(kN/cm) Elastic Displacement(cm) Stiffness(kN/cm) 1.01

3rd Floor 4.78 3F 27.6 1.74 Friction Force(kN) 0.692

2nd Floor 4.78 2F 28.4 3.00 Damping Coefficient(kNs/cm) 0.0333

1st Floor 6.00 1F 35.4 2.27 Damping Ratio 0.037

Total 20.23 Natural Period(sec) 2.84

338
Earthquake Recording in Buildings as a Tool

for the Comprehensive Structural Performance Evaluation

by 12 1 1 Izuru Okawa , Toshifumi Fukuta , Masanori Iiba , Masaomi Teshigawara , Toshihide Kashima2

, Kouichi Kusunoki1, and Shin Koyama3

ABSTRACT

The Building Research Institute (BRI) has been in

charge of the strong motion recordings in

buildings since 1955. The BRI is currently

maintaining totally 78 recording sites nationwide.

After the 1995 Hyogo-ken Nambu earthquake,

recognizing the importance of quick grasp of

damage extent and setup of evacuation and

restoration activity, nationwide strong motion

observation networks such as Kyoshin net (K-net)

maintained by the National Research Institute of

Earth Science and Disaster Prevention, the

Science and Technology Agency (STA) and the

Seismic Intensity seismometers maintained by the

JMA and most of local governments were

established. Massive data are released open to the

public, once big earthquake occurs. In October 6,

2000 and March 24, 2001, large earthquakes

occurred in the western part of Japan. The

previously mentioned networks collected the

ground motion data for wide area surrounding the

epicenters. Those data has actually brought great

advances in the strong motion prediction

methodology. However, when we focus on the

strong motion property variations within the local area or even around one construction site, the

aforementioned network does not necessarily give

sufficient data for evaluating local site effect or

effective input motions for specific building

structures. The instrumentation density is still quite sparse except for large cities. Recently, the

Building Standard Law of Japan has been revised

in accordance with the performance-based-design

oriented scheme. In the scheme, the structural

designer is requested to quantitatively evaluate the

structural performances for various types of

external disturbances including earthquake

motions. The developments of detailed evaluation

techniques for structural performance will be one

of the most important tasks of the BRI. This paper

describes how we can develop and utilize our

strong motion recording system for realizing the

quantitative evaluation of structural performance

against earthquake motions.

KEYWORDS: Earthquake Recording, Structural

Performance, Nationwide Network, Effective

Input Motion

1. INTRODUCTION

The Building Research Institute (BRI) published

many epoch-making records such as the

Kawagishi-cho apartment building, during 1964

Niigata earthquake, recording the tilting action of

the apartment building due to soil liquefaction.

We believe this was the first recording of

liquefaction phenomenon in the world. After that,

we had also obtained large earthquake record in

the campus of Tohoku university, during 1978

Miyagiken-oki earthquake. The earthquake was

recorded at the 1 st and 9th floor of the building on

the campus. The amplitude in the first floor was

still large enough at the time. Attention was drawn

from many researchers that the 9th floor

accelerogram showed amplitude more than 1G.

We had also large amplitude records during 1993

Kushiro-oki earthquake, and 1994 Hokkaido

Toho-oki earthquake and 1994 Sanriku

Haruka-oki earthquake from the recording system.

The location of the recording sites maintained by

BRI is shown in Figure 1 for nationwide recording

sites. The BRI maintains totally 78 recording sites.

There are four types of recording system, i.e.,

(1) Nationwide Strong Motion Recoding Sites

(2) Sendai Dense Instrumental Array in Ground

(3) Recording Network in Tokyo Metropolitan

Area

1 Structural Engineering Department, BRI

2 USEE, BRI

3 National Institute for Land and Infrastructure Management

339
(4) Tsukuba Dense Instrumental Array in Building

and Ground

The detail for each recording system is introduced

in the references. The recording system (1) has the

longest history among the BRI system and has

been continued more than 40 years. The

instrumentation is mostly made in the building.

However, the newer sites has nearby free field

seismometer, if it was made possible, to obtain the

data on effective input motion. Three

instrumentations in the newly built base-isolated

buildings have recently been added.

The recording system (2) in Sendai area is conducted to clarify the soil amplification

characteristics of the various surface soil

conditions. This system does not include the

instrumentation in buildings.

The systems (3) and (4) were completed quite

recently. Therefore, the number of data is not so

large until now. Most of the recording sites

classified into (3) include the recording point

nearby free field. The recording system (4) is deployed on the BRI campus in Tsukuba. It is equipped with 66 channels of sensors for

detecting more detailed behavior of the

building-soil system during earthquake.

2. ROLE OF THE NATIONWIDE STRONG

MOTION OBSERAVATION SYSTEM

After the 1995 Hyogo-ken Nambu earthquake,

recognizing the importance of quick grasp of

damage extent and setup of evacuation and

restoration activity, nationwide strong motion

observation networks such as Kyoshin net (K-net)

maintained by the National Research Institute of

Earth Science and Disaster Prevention, the

Science and Technology Agency (then STA) and

the Seismic Intensity seismometers maintained by

the JMA and most of local governments were

established. Massive data are released open to the

public, once big earthquake occurs. The location of such nationwide recording sites is shown in Figure 2 for only Kanto area. The open

circles indicate the K-net sites and the open

triangles indicate the JMA sites. The solid circles

indicate the BRI sites in the area. In addition to

this, the small dots indicate the location of local

governments in the area. It can be said that large

proportion of the local governments installed

seismometers in their main office. The installation

was not, however, confirmed within the work of

this paper.

In October 6, 2000 and March 24, 2001, large

earthquakes occurred in the western part of Japan.

The previously mentioned networks collected the

ground motion data covering wide area surrounding each of the epicenters.

Figure 3 shows the location of the recording sites

in the central part of Yonago city located in the

western part of Japan facing to the Japan Sea. The

BRI maintains a seismometer on the basement of

Yonago city hall building. The K-net and JMA

also keep the recording site near the BRI site. During the October 6, 2000 earthquake, three sites

obtained large acceleration records. The velocity

response spectra for those motions are shown in

Figure 4. It is seen that there are big difference in

spectral ordinates in spite that the sites locate

closely. This is considered to be due to the

difference of soil conditions and installation

conditions. One should see the spatial distribution

of ground motions to assign which motion

represents the typical motion of the area. Therefore, it can be said that we should see the

severity of ground motion by not only single

recording point data but a scatter of motions

within the area concerned.

3. IMPORTANCE OF EFFECTIVE INPUT

MOTIONS FOR DESIGN PURPOSES

The BRI seismometers had generally been

installed only inside of buildings with few

exceptions since its beginning. Considering the

larger amplitude recorded on the free field sites

compared with the record in building, we came up

to conclude that we need to identify the difference

between the free field and in-building recorded

motions. Recently we set up two series of

earthquake recording systems previously

identified as (3) and (4). One is a system with 2 to 4 three-component sensors in one site installing

on first or basement floor, top floor and nearby

free filed. We installed seismometers at 19 sites

around greater Tokyo metropolitan area. The

recording sites are also shown in Figure 2. It is not

easy, however, to find places for free field, since

there are few free grounds with negligible

influence of buildings, especially in Tokyo. Here,

we basically adopt the selection of the

instrumentation locations in each building as shown in Figure 5. The other is the dense instrumentation to building

and surrounding ground including deep boreholes.

We set up this system in newly constructed

8-story steel reinforced concrete (SRC) building

340
next to the main building on our campus. The

deployment of instruments is shown in Figure 6.

Totally 66 channels will record an event when the

motion exceeds a preset amplitude level. We

believe the recording system of this kind is necessary as a prototype for evaluating the

near-actual behavior of building-soil system,

although this is never a proposal for general

recording system for buildings. The average

spectral ratios of basement and 8

th floor with

respect to the ground surface are shown in Figure

7. Using the sway and rocking model, the

interaction effect is evaluated for this building.

(Kashima, 2001)

4. THE STRATEGIES OF THE EARTHQUAKE

RECORDINGS IN BUILDINGS

Recently we set up the following items as our main target of the strong motion instrumentations.

(1) Evaluation of effective input motions

(2) Detailed database construction indicating

Structural Performance

(3) Promoting the instrumentation in buildings

When we focus our attention on the strong motion

property variations within the local area or even

around one construction site, the aforementioned

nationwide network does not necessarily give

sufficient data for evaluating local site effect or

effective input motions for specific building

structures. The instrumentation density is still quite sparse except for specific large cities.

However, it is not easy to increase the number of

instrumentations only by appealing the necessity

of the earthquake data for buildings. There should

be incentives for promoting the instrumentations

in buildings.

Meanwhile, the Building Standard Law of Japan

has been recently revised in accordance with the

performance-based-design oriented scheme. In the

scheme, the structural designer is requested to quantitatively evaluate the structural performances

for various types of external disturbances

including earthquake motions. The developments

of detailed evaluation techniques for structural

performance should be done and it will become

one of the most important tasks of the BRI. We

need firstly identify that the earthquake recording

gives a reliable tool for evaluating the seismic

performance of buildings. The earthquake

recording is generally still costly. If earthquake

recording is to be utilized for detecting the

vibratory characteristics of building as one of the

crucial performance indices, a cost reduction of

instrumentation will be secondly an important

issue for wide use. When such situation is realized,

more dense locations of instrumentation sites will

be become possible. (See Figure 8)

5. CONCLUSIONS

The nationwide strong motion recording operated

by NIED or JMA are mainly set up for ground

behavior and does not assume some specific

structures. The behavior of each structural system,

such as building, bridge, dam, and port facilities is investigated by the organization as specific

research objectives. A coordination may be

necessary between such organizations with

general and specific purposes for the efficient

utilization of strong motion recording systems. In

addition, the accumulation of the recorded

motions with various classes of buildings

including behavior not only for superstructure but

also foundation and pile foundation as well as ground is our next issue to establish the advanced

structural design schemes in the future. In addition,

the possibility for the earthquake recording to be

used for evaluating the structural performance

should be considered. A low cost instrumentation

and the performance evaluation utilizing the

recorded motion in conjunction with development

of the new design scheme will increase the

database of earthquake response of buildings.

REFERENCES:

1) T. Kashima et al., "Strong-Motion Observation

Program in the Metropolitan Area", (in Japanese)

Proceedings of the AD Annual Meeting, Vol.B-2,

pp.275-276, Kinki, 1997

2) T. Kashima et.al., "Earthquake Motion

Observation in and around 8-Story SRC Building",

(in Japanese) Proceedings of the AU Annual

Meeting, Vol.B-2, pp.2 13-2 14, Kyushu, 1998

3) I.Okawa, et.al., "Dense Instrumentation in BRI

Building and Surrounding Ground", Proceedings

UJNR Workshop on Soil-Structure Interaction,

Menlo Park, California, September 22-23, 1998

4) T. Kashima, "Earthquake Motion Observation

and SSI Characteristics of an 8-story Building in

BRI", The Proceedings of the 2

nd U.S.-Japan

Workshop on Soil-Structure Interaction, March

6-8, Tsukuba, Japan, 2001

5) I. Okawa, "Earthquake Recording for SSI

Study by BRI", The Proceedings of the 2

nd

U.S.-Japan Workshop on Soil-Structure

Interaction, March 6-8, Tsukuba, Japan, 2001

341
Table 1 The list of Recording Sites in Nationwide Strong Motion Observation Network by BRI

Code h I _ Name Instrument Structure SensorPlaces

KGC Kushiro Government Collective SMAC-MDU RC(BI)/B1F+9F+P1F GL,G10, G34, B1F,01F, 09

HRO Hiroo Town Office SMAC-MD RC/2F OIF

HKU Hokkaido University SMAC-MD GL

HKD Hakodate Development and SMAC-MD GL j Construction Department, Hokkaido

Dpvp.lnnmp.nt Riirp.an

HCN Hachinohe City Hall SMAC-MD RC/B1F+6F+P1F B1F,06F

; HCN2 Annex, Hachinohe City Hall SMAC-MDU SRC(BI)/B1F+10F+P1F GL, G30, G105, 10F, OIF, PI ATG Atago Junior High School, Miyako SMAC-MD RC/2F OIF

AKT Akita Prefectural Office SMAC-MD RC/B 1 F+6F+P3F 08F, B1F

THU Tohoku University SMAC-MD SRC/9F 01F,09F

SND Sendai Government Collective Office SMAC-MD S/B2F+15F+P2F B2F, 15F, G40

IWK Iwaki City Hall SMAC-MD SRC/B 1 F+8F+P1

F

B1F,09F

MNM Minamisuna Apartment #3 SMAC-MD SRC7/14F+P3F 15F, OIF

CGC Central Government Collective SMAC-MD S/B3F+20F+P1F OIF, 20B, 19C

OTM Otemachi Government Collective SMAC-MD SRC+S/B3F+15F+P2 16F, B3F

Office #3

BRI Building Research Institute SMAC-MD RC/1F OIF

NIG Niigata City Hall SMAC-MD RC/B 1 F+6F+P2F B1F,07F

JET Joetsu Social Education Office SMAC-MD RC/2F OIF

NGN Nagano Prefectural Office KSG 7/B1F+10F+P2F B1F

ISK Ishikawa Prefectural Office SMAC-MD RC7/B2F+4F+P1F 05F, B2F

KSO Kiso Office, Nagano Prefecture KSG RC/B 1 F+5F+P2F B1F

SMS Shimoda Office, Shizuoka Prefecture SMAC-MD GL

SMK Shimoda-kita High School SMAC-MD RC/4F OIF

HMO Hamaoka Nuclear Power Plant SMAC-MD GL

MTS Matsusaka Office, Mie Prefecture SMAC-MD RC/6F+P1F 07F, OIF

SNG Shingu City Hall KSG RC/B1F+4F+P1F B1F

MIZ Maizuru City Hall KSG RC/4F+P2F OIF, 05F

OSK Osaka Government Collective Office SMAC-MD S/B3F+15F+P3F 18F, B3F

YNG Yonago City Hall SMAC-MD RC/B1F+5F+P1F B1F

HRS Hiroshima Government Collective SMAC-MD SRC/B 1+11F+P2F B1F, 1 IF

Office #2

HMD Hamada Office, Shimane Prefecture SMAC-MD RC/3F+B1F OIF

KCK Kouchi Work Office, Shikoku KSG RC/4F OIF, 04F

Resional Construction Bureau

MYZ Miyazaki Prefectural Office SMAC-MD SRC/B 1F+9F+P1F OIF, 09F

OIT Oita City Hall SMAC-MD SRC7/B2F+9F+P1F OIF, 09F

NBO Nobeoka Office, Miyazaki Prefecture KSG RC/3F+P1F OIF, 04F

MYK Miyakonojo Office, Miyazaki KSG RC/B1F+3F+P2 B1F

YKH Yokohama Government Collective SMAC-MD S/B3F+22F+P2F 23F, B2F

Office #2

HRH Hirosaki Legal Affairs Office SMAC-MDU RC/3F OIF

TRO Tsuruoka Government Collective Offic SMAC-MDU RC/4F+B 1

F

OIF, 04F

SMZ Shimizu Government Collective Office SMAC-MDU SRC/6F OIF, 06F

INvj I Nagoya Government Collective Office oMAL-MUU (jL, JoZr, izr

1 KM Takamatsu Regional Taxation Bureau SMAC-MDU SRC/sr+Blr GL, B1F, 09F

FKO Fukuoka Government Collective Offic< SMAC-MDU SRC/1 0+B1 GL, B1F, 10F

NMW National Museum of Western Art SMAC-MDU RC(BI)/B1F+3F GL, PITE, PITW, 01FE, OIF

SGS Sengen Shrine SMAC-MDU GL

KYM Kiyomuzuyama Park SMAC-MD2 OIF

KSG Koshigaya Branch, Urawa Legal SMAC-MDU IF, 1F[V]

Affairs Bureau

ANX Annex, Building Research Institute AJE-8200 SRC/8F+B1F A01, A14, A43, A89, N14, Bi C01, BFE, BFN, BFS, 1FE, | 2FE, 2FW, 5FE, 5FW, 8FE, 8 8FS, MBC, M5C, M8C, M8E

342
« Observation Sites

Specific Sites in Sendai and

Metropolitan Area

o r \

.Sapporo / Kushlro

1 -.Hakodate V

Hlrosakl n-;

Akita^

Mlyako

Tsuruoka^

^ilg^la

/

^ Sendai Dense Instrument Array

v. ^Sendai (2 sites)

^|waki

K<*gz**a •Nagano Tsultuba / Northern Kantoa '

^Maj|jjru »Wso /

J*'

1

"•^Nagoya Shlmlzu^ (""oWiama

v^^fesdLjh*<2 8ttes>

okyo (3 sites) #.— Network in Tokyo Metropolitan Area

Fukuokji

Hiroshima.. sv

\ ^hlngu

57

'Nobeoka

llyazakl iyakonojo 100 200 400 600 km

Strong Motion Observation Sites operated by Building Research Institute

Figure 1 Strong Motion Recording Sites by

140

140

Figure 2 Strong Motion Recording Sites by BRI (Big solid circle),

K-net (open circle), and JMA (open triangle). Other small dots

indicate the locations of the main office buildings of local governments.

The installation of seismometers thereupon are not confirmed

343
Figure 3 Strong Motion Recording Sites in

Yonago city

Figure 4 Comparison of velocity response

spectra for BRI, K-net and JMA sites

on *ittt»5> ^ *Mrvt-t >-

Hi 6 f£ttttflMRMSK«D-<3»

OK W

Figure 5 Installation of seismometers in buildings/

Upper: Basement and top floors

Lower: Basement, top floor and free field

344
100 m

N14

Loam

Sandy Clay &

Clayey Sand

Sandy Clay

GL-14 m 4 Clay

Fine Sand &

Clayey Sand

A14

Sandy Clay

AClay

GL-43m * A43 Gravel

Sandy Clay

AClay

Gravel &

Fine Sand

Sandy Clay

& Clay

GL-89m

IB Gravel

Figure 6 Dense instrumentation in and around the BRI annex building and surrounding ground

20-

IDS'

Average Fourier Spectral Ratio

- BF/GL (NS)

- BF/GL (EW)

- 8F/GL (NS) -8F/GL(EW)

I

i i 1

1

0.5 1 Frequency (Hz)

20

Figure 7. Average Fourier Spectral Ratio

of records at BF and 8F to GL

Identification of Effective Input Motion

I

Evaluation Techniques for Structural Performance

of Buildings during Earthquakes • Effective Input Motion to Buildings • Detailed Motion Data of Specific Portions of

Buildings • Increase of Database on Structural

Performance • Cost Reduction of Instrumentation for wide use

and software development • Promotion of Performance-based Design

Figure 8. Objectives of Earthquake Recordings in

Buildings

345

APPENDIX

i

Report of Task Committee A

GEOTECHNICAL ENGINEERING AND GROUND MOTION

Date: 13 May 2002

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. Side -- Mehmet Celebi (Co-Chairman) USGS

Michael Sharp (Co-Chairman) CORPS

Harish Chander DOE

Japan Side ~ Osamu Matsuo (Co-Chairman) NILIM

Takahiro Sugano (Co-Chairman) PARI

Keiichi Tamura PWRI

Shojiro Kataoka NILIM

Shuji Tamura Shinshu University

1. Objective and Scope of Work

The objectives of work include:

(1) To promote sharing of strong motion earthquake data among researchers and practicing engineers,

and enhance the availability of technology for evaluating the destructive effects of earthquake

motion, and

(2) To promote research on the dynamic behavior of soils, foundations and earth structures during

earthquakes.

The scope of work includes:

(1) Exchange strong motion data regularly and identify significant issues.

(2) Exchange information on technological developments, state-of-the-art and practice related to

strong motion recording, hazard mapping, soil-structure interaction, soil behavior and stability

during earthquakes,

(3) Exchange field data related to geotechnical engineering, and

(4) Plan and conduct programs of cooperative research and/or workshops in coordination with the

proposed or ongoing programs.

2. Accomplishments

(1) Task committee published and distributed (CD-ROM) the proceedings of the second U.S. -Japan

Workshop on Dynamic Soil-Structure Interaction that was held in Tsukuba, Japan, March 6-

8,2001.

(2) Through US side secretary general and through Japanese side experts, a new CD-ROM data set containing the main-shock and aftershock records of the 1999 Izmit and Duzce, Turkey

earthquakes (USGS OFR 01-163) was provided to Japan side.

(3) The Task Committee members have been cooperating in the following research projects:

a) Physical and Numerical Simulation of Structural Damages Due to Liquefaction and

Development of Countermeasure Techniques, 1994-2004, by the National Research

Institute for Earth Science and Disaster Prevention (NIED), Building Research Institute

(BRI) and Wayne State University (WSU), and

b) Application of the Nuclear Magnetic Resonance Imaging (NMRI) Method to the Study of

Soil Behavior and Stability During Earthquakes, by Port and Airport Research Institute

(PARI) and Colorado School of Mines and the Lovelace Institute.

349
3. Future Plans

The future plans include:

(1) Plan to hold the fifth workshop on geotechnical earthquake engineering in early 2003 in San

Diego, in conjunction with the workshop on seismic performance of urban, reclaimed and port

areas. The workshop will be coordinated by Dr. T. Sugano and Dr. Michael Sharp-co-chairmen of

the Task Committee and Prof. Scott Ashford of U.C., San Diego.

(2) Plan the 3

rd US-Japan Workshop on Soil-Structure Interaction for 2003 {Responsible members:

Celebi & Okawa).

(3) Plan a workshop on Real-Time Instrumentation of Densely Urbanized Areas and Instrumentation

of Civil Engineering Structure. [Temporary Committee: Fujiwara, Okawa, Kusakabe, Borcherdt,

Wald, Celebi]

(4) Continue the cooperative research "Physical and Numerical Simulation of Structural Damages

Due to Liquefaction and Development of Countermeasure Techniques, 1994-2004", by the

National Research Institute for Earth Science and Disaster Prevention (NIED), Building Research

Institute (BRI) and Wayne State University (WSU). [Responsible Members: Dr. Minowa and

Professor Kagawa].

(5) Explore the possibility of joint research on site response issues (Responsible Members: Dr.

Matsuo & Dr. Borcherdt).

350
Report of Task Committee B

NEXT-GENERATION BUILDING AND INFRASTRUCTURE SYSTEMS

Date: 13 May 2002

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. Side — George Lee (Co-Chairman) MCEER

Peter Chang (Co-Chairman) NSF

H.S. Lew NIST

Adam Rose Penn Sate

William Roper GWU

Josephine Malilay CDC

Japan Side — Masaomi Teshigawara (Co-Chairman) BRI

Keiichi Ohtani (Co-Chairman) NIED

Kazuhiko Fujihashi NTT

1. Objectives and Scope of Work

1) Objective

The objective of the task committee is to enhance the development and implementation of innovative

and new (i) materials; (ii) enabling technologies; (iii) evaluation, analysis, design, construction and

maintenance methods; through cooperated individual and organized and networked analytical and

experimental approaches for the next - generation building and infrastructure systems.

Opportunities during the next few years include experimental observations on the performance of

model and near full - size structures using large scale testing facilities and networked earthquake

engineering experimental facilities, and the development of innovative technologies and approaches to address the newly obtained analytical conclusions and observed experimental evidences and to formulate various performance based engineering guidelines for building and infrastructure systems.

2) Scope of work

(1) Enhancement of exchange of information and personnel

(2) Coordinate joint research including the utilizing experimental facilities.

(3) Coordinated development of database, test procedures, guidelines for test result interpretation

and application.

(4) Development of new materials and technologies for condition assessment, retrofit of existing

buildings and design of new buildings.

(5) Conduct joint workshops and joint meetings to identify research opportunities, new projects, and

make recommendation for UJNR Panel.

2. Accomplishments

(1) The US-Japan grantees' meeting for Composite and Hybrid Structures, June 24 to 27, 2001 in

Berkeley, CA. The committee will plan for the publication of CHS design guides and other

research reports summarizing the experimental and analytical research activities.

(2) A workshop to identify and develop specific joint projects and cooperative activities within

the general scope of Task Committee B was held at NSF attended by the four co-chairs

together with Dr. S.C. Liu in August 13 of 2001.

(3) In February 27 of 2002, the two co-chairs of US and Japan together with Dr. S.C. Liu

coordinated the session program of UJNR meeting.

351
3. Future Plans

(1) The 3rd U.S. -Japan Technical Coordinating Committee Meeting on Auto-Adaptive Media (Smart

Structural Systems) will be held the end of October 2002, in Tsukuba, Japan.

(2) Task Committee B reviewed the status report of NEES (Network of Earthquake Engineering

Simulation) provide by Dr. Joy Pauschke of NSF and expresses its desire for JTCC and/or other

appropriate US and Japan groups to organize meetings to establish a US-Japan Coordinating

research plan that will accelerate NEES program in US and performance-based design focusing on

building functions, and health monitoring.

(3) Task Committee B plans to include societal and economic impact analysis in future activities, with

initial emphases given to information technology, public health and security aspects of

development of next generation building and infrastructure system. Alternatively, because

information technology, infrastructure security, and disaster-related public health issues are important to activities of other task committees, it may be desirable to establish a separate

panel-wide (task) committee to address such "trans-panel" or ''trans-committee" activities as needs

and opportunities arise.

4. Related Activity

(1) The 3rd grantee meeting, as part of the US-Japan cooperative research on urban earthquake disaster

mitigation workshop was held on August 15-16, 2001, Seattle, Washington.

(2) JTCC for the discussion on next US/Japan cooperative research was held after the 3rd grantee

meeting.

(3) The 4th grantee meeting, as part of the US-Japan cooperative research on urban earthquake disaster

mitigation workshop will be held on October 21-22, 2002, Kyoto, Japan.

(4) The foundation work of the 3-D Full-scale Earthquake Testing Facility at Miki by NIED has been

completed. Super structures are under construction. And whole facility will be completed by early

2005.

352
Report of Task Committee C

DAMS

Date: 13 May 2002

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. Side - Dr. Robert L. Hall (Chairman)

Dr. Enrique E. Matheu

USACE-ERDC

LSU

Japan Side Dr. Yoshikazu Yamaguchi (Chairman)

Dr. Ken-ichi Torii

PWRI

NILIM

1. Objectives and Scope of Work

To develop technical insights into better understanding of the response of dams to seismic effects, the T/C

will plan, promote, and develop research initiatives to assist in assuring seismic safety and economical

protective countermeasures against earthquake loading for these critical structures.

The scope of work includes:

(1) Develop methods of analysis for seismic design of dams including outlet works:

- Comparative analysis of design methods and evaluation criteria used by U.S. and Japan. - Development of "Design Earthquake Ground Motions" for analysis and evaluation of dams. - Assessment of models and numerical procedures used for seismic dynamic analysis.

(2) Perform research to better understand the dynamic characteristics of dam construction materials

and site conditions.

- Strength and deformation characteristics of concrete, soil and rock materials during

earthquakes.

(3) Evaluate observed performance of dams and outlet works during earthquakes.

- Develop a database that contains measured ground accelerations and dynamic response of

dams and outlet works during earthquakes, and other related information necessary to

evaluate their seismic behavior, such as experimental and/or analytical data with description

of test and analysis procedures used. - Investigation of damage mechanisms due to earthquake loading. - Application of the analysis of the observed dynamic behavior to the improvement of design

criteria.

2. Accomplishments

(1) The Task Committee developed a database that contains measured ground accelerations and

dynamic response of dams and outlet works during earthquakes. The database has been expanded

to include other relevant information necessary to evaluate their seismic behavior, such as experimental and/or analytical data with description of test and analysis procedures used.

(2) In investigating seismic design ground motions at dam sites in Japan, technical information from

U.S. in this field was referred to.

353
(3) The U.S.- and Japan-side Task Committee members have actively planned, advertised, and

organized the upcoming "3rd U.S.-Japan Workshop on Advanced Research on Earthquake

Engineeringfor Dams", San Diego, USA, June 22-23, 2002:

- The Organizing Committee is chaired by Dr. Robert Hall (U.S. Army Corps of Engineers),

and it also includes Dr. Yoshikazu Yamaguchi (PWRI), Dr. Enrique E. Matheu (Louisiana

State University) and Dr. Ziyad H. Duron (Harvey Mudd College). - The workshop is co-sponsored by the U.S. Army Engineer Research and Development Center

(ERDC) and the U.S. Society on Dams (USSD). It will be part of the pre-conference

activities of the 22nd USSD Annual Meeting and Conference (to be held at the same location

on June 24-28, 2002).

3. Future Plans

(1) The Task Committee will continue the current efforts focused on the development of improved

mechanisms to facilitate the continuous exchange of results of research activities and general

technical information related to dam earthquake engineering.

(2) The Task Committee will coordinate exchange visits of scientists and engineers from the U.S. and

Japan. A series of technical problems of mutual interest will be identified and prioritized and they

will serve as the focus for this exchange program.

(3) The Task Committee will promote the development of a research program to improve test and

analysis procedures based on the comparative study of representative case studies that will be

available in the jointly developed database

(4) The Task Committee will conduct the "3rd U.S.-Japan Workshop on Advanced Research on

Earthquake Engineering for Dams" June 22-23, 2002. The Task Committee will compile the

technical contributions in the corresponding Workshop Proceedings that will be published by

ERDC later in the year.

- The workshop will focus on a variety of technical aspects covering concrete and embankment

dams, and outlet works. The technical sessions will address seismic evaluation criteria,

numerical techniques for seismic analysis, seismic design and evaluation of appurtenant

structures, instrumentation, condition assessment, seismic upgrade and rehabilitation,

foundation problems, case studies, as well as many other general topics. - The program includes several keynote lectures and invited presentations from wellrecognized participants from government agencies, academia, and private sector. The

technical presentations include eight papers from Japan and 20 papers from the U.S. and

other countries.

4. Task Committee Members:

U.S. Side ~ Dr. Robert L. Hall (Chairman) USACE-ERDC

LSU

USACE

USACE

USACE-ERDC

USACE-ERDC

USACE

PWRI

NILIM

Japan Dam Eng. Ctr.

Dr. Enrique E. Matheu

Ms. Anjuna Chudgar

Mr. David Dollar

Dr. Richard C. Dove

Dr. Richard Olsen

Mr. Rick Peoppelman

Japan Side — Dr. Yoshikazu Yamaguchi (Chairman)

Dr. Osamu Matsuo

Mr. Kaoru Niwa

354
Dr. Choshiro Tamura

Dr. Tatsuo Omachi

Dr. Yoshio One

Mr. Takashi Tokoro

Mr. Shuju Takasu

Dr. Yasuhiko Wakizaka

Dr. Kei-ichi Tamura

Mr. Takashi Sasaki

Mr. Hideaki Kawasaki

Dr. Ken-ichi Torii

Mr. Fumio Yonezaki

Mr. Akira Hara

Nippon University

Tokyo Inst, of Tech.

Aichi Inst, of Tech.

Japan Inst. Civil Eng.

PWRI

PWRI

PWRI

PWRI

NILBVI

NILIM

Water Resources Dev. Public Corp.

Water Resources Dev. Public Corp.

355

Report of Task Committee D

WIND ENGINEERING

Date: 13 May 2002

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. Side - Dr.

Dr.

Dr.

Dr.

Dr.

Dr.

Peter Chang (Chairman) NSF

Partha Sarkar Iowa State U.

Emil Simiu NIST

Fahim Sadek NIST

Nicholas Jones

Harold Bosch

Johns Hopkins

FHWA

Japan Side Dr.

Mr.

Dr.

Dr.

Mr

Hisashi Okada (Co-Chairman)

Jun Murakoshi (Co-Chairman)

Hiroshi Sato

Yasuo Okuda

Koichiro Fumoto

BPJ

PWRI

PWRI

BPJ

PWRI

1. Objective and Scope of Work

To exchange technical information and to jointly plan, promote, and foster research and dissemination, to improve understanding of wind and its effects on structures, establish more rational wind resistant design

methods for structures, and to contribute to wind hazard mitigation. Specific objectives include:

(1) Facilitate collaborative research and cooperation between U.S. and Japanese researchers in

wind engineering.

(2) Identify nuggets of success in wind engineering.

(3) Identify research needs in the areas of hazards mitigation, new technology, basic and applied

research and social and economic impacts of wind hazard.

The scope of work includes:

(1) To characterize strong wind, especially boundary layer extreme winds.

(2) To study wind effects including wind loading on and wind-induced response of structures.

(3) To perform experimental and analytical methods to predict wind and its effects.

(4) To perform damage, wind hazard, and risk assessments and develop technologies for wind

To facilitate the above scope of work, conduct annual or biannual workshops to evaluate current and

future research agendas that may lead to synergistic research between U.S. and Japan. The workshops

should be alternatively held in U.S. and Japan.

2. Accomplishments

(1) Following the Panel's Strategic Plan, approved at its 33

rd Joint Meeting, the former Task

Committee E, Design for Wind and Wind Hazard Mitigation was reactivated as Task Committee

D, Wind Engineering. This decision was based on the Task Committee's past accomplishments

and its solid approach as outlined in its Future Plans below that address its Objective.

hazard mitigation.

357
3. Future Plans

(1) Plan a workshop on "Design for Wind and Wind Hazard Mitigation." This third U.S.-Japan

Workshop will be held in Seattle, U.S. in September 2002. The theme of this Workshop is "Reducing Losses from Wind: Collaborative Opportunities for the 21

st Century."

(2) Pursue collaborative research on the following topics

a. Wind characteristics and wind hazards

i. Wind effects due to complex terrain

ii. Modeling and validation with full-scale data

iii. Joint field studies

iv. Cooperative quick-response post-storm damage assessments

b. Wind effects on buildings

i. Development of realistic aerodynamic loads for performance-based structural design

ii. Comparison of wind tunnel data sets and CFD results to full-scale measurements

c. Wind effects on bridges

i. Improvement of methods for predicting wind-induced response of cable-suspended

bridges with edge girders

ii. Prediction and mitigation of rain/wind-induced vibration of stay cables

d. Evolving Technologies

i. Next generation wind tunnels: Improvement of methods to simulate non-stationary

wind (straight-line, tornado and microburst)

ii. Structural control

(3) Exchange technical information on the following topics

a. Wind characteristics and wind hazards

b. Wind pressures, loadings and dynamic characteristics of buildings

c. Wind-induced response of cable-suspended bridges

d. Prediction and mitigation techniques for wind effects

358
Report of Task Committee G

TRANSPORTATION SYSTEMS

Date: 13 May 2002

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. Side - W. Phillip Yen (Chairman) FHWA

FHWA

CALTRANS

Hamid Ghasemi

Ray Zelinski

Japan Side Hiroshi Sato (Chairman)

Katsuhiko Kawashima

PWRI

TIT

PWRI

PWRI

NILIM

NILIM

Shigeki Unjoh

Keiichi Tamura

Katzuhiro Nishikawa

Ishida Masahiro

1. Objective and Scope of Work

The objectives of work include:

(1) To plan, promote and foster research on the behavior of transportation facilities when subjected to wind and seismic forces, and

(2) To disseminate research results and provide specifications and guidelines based on the Task

Committee's findings.

The scope of work includes:

(1) To investigate existing and new bridges design, the behavior of whole bridge systems and/or

single component of a bridge without limitation on their size and function, and

(2) To investigate offshore, coastal and airport facilities' behavior during earthquake.

2. Accomplishments

(1) The 17th U.S. - Japan Bridge Engineering Workshop was held during 12-14 November 2001, at Tsukuba, Ibaraki, Japan. 16 U.S. and 73 Japanese participants attended the workshop. 47

technical papers were presented and discussed on the following main topics: a) Bridge

Management Technology and b) Evaluation of Seismic Performance and Seismic strengthening.

(2) Dr. Phillip Yen (FHWA) was invited to the PWRI during November 18 -22, 2001 to discuss the

experimental evaluation methods for establishing a Guidelines for Seismic Performance Testing

in Bridge Structures.

(3) Dr. Hiroshi Sato (PWRI) visited the Turner - Fairbank Highway Research Center on March 7,

2002, to discuss future plans of the Task Committee with Dr. Phillip Yen, and to exchange

technical information on bridge management and bridge aerodynamics.

3. Future Plans

(1) The proceedings of the 17th U.S. - Japan Bridge Engineering Workshop will be published soon and distributed by the Japan-side. The 18th U.S. - Japan Bridge Engineering Workshop will be

held in October 2002 in the U.S. Specific dates, location, themes, program, and itinerary will be

proposed by the U.S. -side Task Committee G with the concurrence of the Japan-side Task

Committee.

359
(2) Conduct the following cooperative researches:

a) Seismic Performance Testing of Bridge Structures. This program is to establish a Guidelines for seismic performance testing of bridge structure. The objective of this

program is to establish a Uniform National Guidelines for Testing Bridge Components,

which will increase comparability of test results and reduce the duplications of similar

testing. In 2002, FHWA and PWPJ will meet to develop a draft common Guidelines for

the experimental methods. The 1

st US side Expert Panel Meeting will be held in August

2002.

(3) Following a devastated earthquake or hurricane (typhoon) in the US or Japan, the committee will

form a joint reconnaissance team to investigate the performance of transportation system.

(4) Exchange technical information on the following topics:

a) Seismic design and retrofit of highway bridges including the development of seismic

response modification devices, smart materials such as SMA, seismic retrofit for longspan bridges, and the development of damage evaluation methods using sensor systems.

b) Seismic and aerodynamic response of long span bridges including suspension and cablestayed bridges, with emphasis on behavior of composite materials, cable inspection,

vibration control, and corrosion protection.

(5) Encourage coordinated researches on the following topics:

a) Bridge design methodologies that enhance performance based design concepts, limit state design concepts and life cycle cost analyses.

b) Modeling of seismic isolator and the seismic isolation design methods.

c) System identification techniques, non-destructive evaluation of bridge structures, use and

performance of structural materials including new materials, and performance ofjointless

bridges.

d) Bridge response due to near fault effects.

e) Bridge foundation performance -based design methods.

f) Post-earthquake damage evaluation methods for seismic retrofitted bridges.

4. Others

(1) The 3

rd International Transportation Structure Workshop on Bridge Seismic Performance-Based

Design will be held in July 9-11 at Tokyo Institute of Technology, Tokyo, Japan, Task committee

members are encouraged to participate.

360 !
Technical Publications

Periodical

Journal of Research of the National Institute of Standards and Technology—Reports NIST research

and development in metrology and related fields of physical science, engineering, applied mathematics,

statistics, biotechnology, and information technology. Papers cover a broad range of subjects, with major

emphasis on measurement methodology and the basic technology underlying standardization. Also included

from time to time are survey articles on topics closely related to the Institute's technical and scientific

programs. Issued six times a year.

Nonperiodicals

Monographs—Major contributions to the technical literature on various subjects related to the

Institute's scientific and technical activities. Handbooks—Recommended codes of engineering and industrial practice (including safety codes) developed in cooperation with interested industries, professional organizations, and regulatory bodies.

Special Publications—Include proceedings of conferences sponsored by NIST, NIST annual reports, and

other special publications appropriate to this grouping such as wall charts, pocket cards, and bibliographies.

National Standard Reference Data Series—Provides quantitative data on the physical and chemical

properties of materials, compiled from the world's literature and critically evaluated. Developed under a worldwide program coordinated by NIST under the authority of the National Standard Data Act (Public

Law 90-396). NOTE: The Journal of Physical and Chemical Reference Data (JPCRD) is published

bimonthly for NIST by the American Institute of Physics (AIP). Subscription orders and renewals are

available from AIP, P.O. Box 503284, St. Louis, MO 63150-3284.

Building Science Series—Disseminates technical information developed at the Institute on building

materials, components, systems, and whole structures. The series presents research results, test methods, and

performance criteria related to the structural and environmental functions and the durability and safety

characteristics of building elements and systems.

Technical Notes—Studies or reports which are complete in themselves but restrictive in their treatment of

a subject. Analogous to monographs but not so comprehensive in scope or definitive in treatment of the

subject area. Often serve as a vehicle for final reports of work performed at NIST under the sponsorship of

other government agencies.

Voluntary Product Standards—Developed under procedures published by the Department of Commerce

in Part 10, Title 15, of the Code of Federal Regulations. The standards establish nationally recognized

requirements for products, and provide all concerned interests with a basis for common understanding of

the characteristics of the products. NIST administers this program in support of the efforts of private-sector

standardizing organizations.

Order the following NIST publications—FIPS and NISTIRs—from the National Technical Information

Service, Springfield, VA 22161.

Federal Information Processing Standards Publications (FIPS PUB)—Publications in this series

collectively constitute the Federal Information Processing Standards Register. The Register serves as the

official source of information in the Federal Government regarding standards issued by NIST pursuant to the Federal Property and Administrative Services Act of 1949 as amended, Public Law 89-306 (79 Stat.

1 127), and as implemented by Executive Order 1 1717 (38 FR 12315, dated May 1 1, 1973) and Part 6 of

Title 15 CFR (Code of Federal Regulations).

NIST Interagency or Internal Reports (NISTIR)—The series includes interim or final reports on work

performed by NIST for outside sponsors (both government and nongovernment). In general, initial distribution is handled by the sponsor; public distribution is handled by sales through the National Technical

Information Service, Springfield, VA 22161, in hard copy, electronic media, or microfiche form. NISTIR 's may also report results of NIST projects of transitory or limited interest, including those that will be

published subsequently in more comprehensive form.
oU -acs o 55

ooi Os

OS

00O

oom