Information Circular 9453

Proceedings: New Technology For Coal Mine

Roof Support

U.S. DEPARTMENT OF HEALTH AND HUMAN SERVICES

Public Health Service

Centers for Disease Control and Prevention

National Institute for Occupational Safety and Health

Pittsburgh Research Laboratory

Pittsburgh, PA

October 2000
This document is the public domain and may be freely copied or reprinted.

Disclaimer: Mention of any company or product does not constitute endorsement by NIOSH.

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DHHS (NIOSH) Publication No. 2000-151
CONTENTS

Page

Abstract ...................................................................................... 1

GENERAL SUPPORT CONSIDERATIONS

Roof and rib fall incidents and statistics: a recent profile, by D. M. Pappas, E. R. Bauer, and C. Mark, Ph.D. ...... 3

Fundamentals of coal mine roof support, by C. Mark, Ph.D., and T. M. Barczak . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 23

PRIMARY ROOF SUPPORT

Trends in roof bolt application, by D. R. Dolinar and S. K. Bhatt . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 43

Assessing coal mine roof stability through roof fall analysis, by G. M. Molinda, C. Mark, Ph.D., and D. R. Dolinar . 53

Load behavior of grouted bolts in sedimentary rock, by S. P. Signer . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 73

Summary of field measurements of roof bolt performance, by C. Mark, Ph.D., D. R. Dolinar, and T. M. Mucho . . . . 81

Skin failure of roof and rib and support techniques in underground coal mines, by E. R. Bauer and D. R. Dolinar . . . 99

Design of roof bolt systems, by C. Mark, Ph.D. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 111

SECONDARY ROOF SUPPORT

Design methodology for standing secondary roof support systems, by T. M. Barczak, T. P. Mucho, and

 D. R. Dolinar . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 133

Optimizing secondary roof support with the NIOSH Support Technology Optimization Program (STOP), by

 T. M. Barczak . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 151

Cable support in longwall gate roads, by D. R. Dolinar and L. A. Martin . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 165

Material handling considerations for secondary roof support systems, by T. M. Barczak . . . . . . . . . . . . . . . . . . . . . . . 193

LONGWALL SHIELDS

NIOSH safety performance testing protocols for standing roof supports and longwall shields, by T. M. Barczak . . . . 207

Examining longwall shield failures from an engineering design and operational perspective, by T. M. Barczak . . . . . 223

Factors to consider when purchasing a new set of longwall shields, by T. M. Barczak . . . . . . . . . . . . . . . . . . . . . . . . . 245

GROUND CONTROL INSTRUMENTATION

Instruments for monitoring stability of underground openings, by M. K. Larson, D. R. Tesarik, J. B. Seymour, and

 R. L. Rains . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 253

Evaluation of ground support at a trona mine using instrumented cable and rebar bolts, by L. A. Martin,

 C. A. Dreesbach, and R. Pakalnis, Ph.D. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 271
iii

UNIT OF MEASURE ABBREVIATIONS USED IN THIS REPORT 

cm centimeter ksi kips per inch

cm2 square centimeter kS kilohm

ft foot lb pound

ft2 square foot lb/d pound per day

ft3 cubic foot lbf pound of force

gal/min gallon per minute m meter

GPa gigapascal m2 square meter

hr hour m3 cubic meter

in inch min minute

in2 square inch mm millimeter

in3 cubic inch MPa megapascal

kcal/min kilocalories per minute N newton

kg kilogram psf pound per square foot

KHz kilohertz psi pound per square inch

kN kilonewton sec second

kN/d kilonewton per day V volt

kPa kilopascal
PROCEEDINGS: NEW TECHNOLOGY FOR COAL MINE ROOF SUPPORT

Edited by Christopher Mark, Ph.D.,1 Dennis R. Dolinar,2 Robert J. Tuchman,3

Thomas M. Barczak,4

 Stephen P. Signer5 and Priscilla F. Wopat6

ABSTRACT

Roof falls continue to be the greatest single safety hazard faced by underground coal miners. During 1996-

99, 44 coal miners lost their lives in rock falls, and nearly 2,400 were injured. In addition, nearly

6,000 noninjury roof collapses were reported. Roof supports are installed to protect the miners, but support

system failures contributed to most of these incidents.

Reducing the terrible toll taken by ground falls continues to be a major goal of research by the National

Institute for Occupational Safety and Health (NIOSH). The purpose of these proceedings is to provide the

mining community with a comprehensive survey of coal mine roof supports. Drawing on many years of

research undertaken by the NIOSH Pittsburgh and Spokane Research Laboratories, this volume describes what

types of support are available, how they work, and when they should be used. The major subjects covered

include roof bolts, standing roof supports, cable supports, and longwall shields. Some special topics are also

addressed, including an analysis of roof fall accident statistics, techniques for better skin control, material

handling considerations, and longwall mining through recovery rooms.

This proceedings volume also contains information on several important new technologies, which are

described here for the first time:

$ Guidelines for selecting roof bolt length, pattern, and capacity that were derived from statistical analysis

of the roof fall experience at 37 underground mines;

$ A new design method for longwall tailgate supports; and

$ A technique for measuring loads developed within cable bolts.

The papers in these proceedings were presented at open industry briefings conducted by NIOSH on New

Technology for Coal Mine Roof Support. The briefings were held in Norton, VA, Charleston, WV,

Evansville, IN, Tuscaloosa, AL, Price, UT, Glenwood Springs, CO, and Washington, PA. The papers were

also presented at the Mine Safety and Health Administration (MSHA) Preventative Roof-Rib Outreach

Program (PROP) Seminar, which was held at the National Mine Health and Safety Academy near Beckley,

WV.

1

Supervisory physical scientist, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA. 2

Mining engineer, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.

3

Technical writer-editor, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.

4

Research physicist, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.

5

Mining engineer, Spokane Research Laboratory, National Institute for Occupational Safety and Health, Spokane, WA.

6

Technical publications editor, Spokane Research Laboratory, National Institute for Occupational Safety and Health, Spokane, WA.
3

ROOF AND RIB FALL INCIDENTS AND STATISTICS:

 A RECENT PROFILE

By Deno M. Pappas,1 Eric R. Bauer,2 and Christopher Mark, Ph.D.3

ABSTRACT

During 1998-99, groundfall incidents resulted in 27 fatalities and were responsible for over 70% of all

deaths in U.S. underground coal mines. To obtain a better understanding of where and why these incidents

occurred, a comprehensive analysis of groundfall injuries and fatalities was conducted. The first portion of

the study examined various factors associated with roof and rib fall injuries and reportable roof fall noninjuries

that occurred during 1995-98. The study found that the room-and-pillar mining method has twice the

groundfall incident rate than the longwall method. Mine locations with high groundfall rates seem to correlate

to regions where there is a higher concentration of problematic coalbeds. For example, the Illinois Basin has

very high groundfall rates, which can be traced back to several key coalbeds-Kentucky No. 13, Herrin/No. 6/

Kentucky No. 11, and Springfield No. 5/Kentucky No. 9. High rib fall rates were found in mines located in

thick seams. Groundfall rates were found to be 30% to 40% higher during the months of July through

September, possibly due to high humidity that may cause the shale mine roof to deteriorate.

The second part of the study examined the root causes of failure by reviewing all groundfall fatality reports

for 1996-99. Primary and secondary hazard factors were assigned to each groundfall incident. The primary

factors resulting in these groundfall fatalities were pillar extraction, traveling under unsupported roof, skin

failure, construction, longwall faces, intersections, and geologic discontinuities. Defining prominent ground

control incident trends and hazards will identify areas where additional study is needed and where innovative

solutions need to be developed to reduce these severe occupational hazards.

1

Civil engineer.

2

Mining engineer.

3

Supervisory physical scientist.

Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
4

Figure 1.—Historical overview of groundfall fatalities, 1910-98. (Source: MSHA data.)

INTRODUCTION

Underground coal mining has always been recognized as one

of the most hazardous occupations in the United States. Since

1910, more than 85,000 underground miners have lost their

lives while mining coal. Approximately 47% of these fatalities,

involving 40,000 miners, have occurred because of falls of roof

and rib, which is a greater proportion than for any other type of

incident classification. As shown in figure 1, more than 1,000

roof and rib fall fatalities occurred annually until the early

1930s. Decreases in roof and rib fatalities in the 1930s and late

1940s seem to coincide with drops in production due to

slowdowns in the economy during to the Great Depression and

the post-World War II recession. The use of roof bolts and the

mechanization of the mining industry in the early 1950s

considerably improved productivity and required a smaller

workforce. As the number of miners worked underground

decreased, so did the number of groundfall fatalities. However,

the actual rate of roof and rib fatalities did not significantly drop

until after the early 1960s. This drop may be due to several

factors, including greater use of mechanical and then resingrouted roof bolts, the Federal Coal Mine Health and Safety

Act 

of 1969, research by the former U.S.Bureau of Mines (USBM)

in the development of automated temporary roof supports

(ATRS) and canopies, and the creation of the Mine Safety and

Health Administration (MSHA), which mandated ground

control safety and training programs [Pappas 1987].

Although the frequency of roof and rib fatalities has

significantly decreased, from 1,300 fatalities in 1910 to 14

fatalities in 1998, groundfalls still injured 850 workers in 1998,

resulting in 26,000 days lost. In addition to the injuries, more

than 1,800 noninjury roof falls occurred in 1998. These

noninjury roof falls were usually massive falls that extended

beyond the height of the bolts, damaged equipment, stopped

production, or disrupted ventilation. It has been estimated that

the total cost of groundfall injuries during 1985-89 was $123.9

million, or approximately $1 million for every fatality and

$6,835 for every injury [Peters and Randolph 1991].

The USBM identified increases and/or patterns that

influenced groundfall injury rates during 1980-84 associated

with mining method, mining height, geographic location, and

mine size [Pappas 1987]. This study found that during that
5

period, small mines (<50 employees) had severe groundfall

fatality incidence rates about 5 to 20 times the fatality rate of

medium- and large-sized mines. A slightly different study

conducted by Fotta et al. [1995] found that in the Appalachian

region in the early 1990s, small-mine groundfall fatality rates

were elevated, but were significantly decreasing from levels in

previous years. Although these studies are not identical, they

indicate that dramatic changes were occurring in underground

coal mines that were dynamically influencing groundfall

patterns and characteristics. A major factor that may have

influenced these fatality rate changes is the dramatic rise in use

of the longwall mining method. Longwall mining in the United

States has revolutionized and streamlined the process of mining

underground coal by vastly improving mine productivity and

miner safety. When the USBM study was conducted in 1984,

longwall mining accounted for about 18% of the underground

production; by 1994, however, longwall production doubled to

about 37%. These more efficient longwall mines may have

resulted in many smaller room-and-pillar mines closing or

becoming more productive and safer in order to stay

competitive with the longwall mines. This may have resulted

in the dramatic decrease in small-mine fatality rates. Consequently, the goal of this study is to update groundfall fatality,

injury, and noninjury incidence rates where the significant

factor of longwall mining can be evaluated separately from

room-and-pillar mining operations. The various attributes

examined include mining method, location, mine size, seam

height, coalbed, and seasonal patterns. In addition, specific

hazards contributing to groundfall fatalities are discussed based

on details compiled from MSHA fatality reports.

METHODOLOGY

Raw data for this study were obtained from the MSHA

accident and address (listing of all operating mines) databases.

The study examined the close-out data for the period 1995-98

for underground coal mines, excluding contractors. Ground

control incidents included all roof and rib falls listed in the

database, as well as incidents classified as "machinery" where

the source of injury was caving rock. The narratives of these

machinery-related groundfalls were reviewed to classify

whether the incident was a roof or rib fall. For this study, roof

and rib fall incidents were categorized into four groups:

1. Roof fall injury incidents resulting in death, in a

permanent disability, in days away from work or days of

restricted work activity, or in no lost workdays or restricted

activity (degree of injury: 1-6).

2. Noninjury reportable roof fall incident (degree of

injury: 0). 30 CFR 50.20-5 requires that every roof fall in

active workings that occurs above the roof bolt anchorage,

impairs ventilation, or impedes passage be reported.

3. Fatal roof or rib fall incident (degree of injury: 1).

4. Rib fall injury incident resulting in the same injuries

listed in item 1 above.

The noninjury reportable roof fall incidence rate is important

because it consistently tracks catastrophic roof failures

according to MSHA regulations. A high noninjury roof fall rate

may indicate situations where severe roof control problems are

occurring since these incidents are associated with massive roof

failure.

All incidents are converted into an incidence rate by

computing the number of roof or rib fall incidents divided by

the number of hours worked underground per 200,000 hr. The

200,000 hr approximates the number of hours worked by

100 full-time miners per year.

The two methods of mining-room-and-pillar versus

longwall-were separated out for most of the analyses. The

longwall mines were identified for each year by review of the

annual longwall census and knowledgeable longwall mining

individuals. It was determined that most longwall mines are

dedicated exclusively to mining longwalls and gate road

development; these mines were not proportioned by mining

method. Therefore, all production originating from these

designated mines were determined to be longwall mines;

everything else was designated as room-and-pillar mine.

Major coal-producing locations in the United States were

examined to identify trends. The eastern Kentucky region has

minimal longwall production; by the end of the study period, no

longwall was operational. Although the eastern Kentucky rates

are listed, they should not be considered in the study since they

represent very little production. Eastern Pennsylvania's

anthracite coalfields are rather unique due to their folded

stratigraphy, which is more similar to hard-rock mines than

bituminous coal mines. Ground control conditions in eastern

Pennsylvania are not equivalent to the other coal-mining

locations, but are listed for consistency purposes.

Seam height used in the MSHA database is defined as the

average height of coal seam currently being mined. There may

be some variation in the seam height at the actual site of the

incident versus the seam height listed in the MSHA database.

Although the fatality reports record the actual working seam

height at the site of the incident, this information is not

available for all of the other types of injuries. To be consistent,

this study will use the MSHA database seam height. Three

categories of seam height are usually selected based on

distribution of employee hours and other constraints [Fotta and

Mallett 1997]. Thin-seam heights are #42 in, medium-seam

heights are 43-60 in, and thick-seam heights are $61 in.

Approximately 1.3% of the total underground hours are from
6

mines where no seam height was recorded or was misreported

and so were excluded from this portion of the study.

Mine size is based on the average annual number of

employees working in the underground mine. Three categories

of mine size that were selected are based on distribution of

employee hours. Small-sized mines have #49 workers,

medium-sized mines have 50 to 149 workers, and large-sized

mines have $150 workers.

To obtain a greater understanding of the specific hazards

associated with roof or rib fall fatalities, all underground

groundfall fatality reports were examined for 1996-99. The

fatality reports provide much more information than can be

obtained from the MSHA injury narratives. The cause of

failure for each groundfall fatality was reviewed by several

individuals to minimize subjectivity and were categorized into

eight primary and secondary hazard groups.

MINE CHARACTERISTICS

Table 1 breaks out several characteristics associated with

room-and-pillar coal mines, such as the mining method,

location, mine size, and seam height. Table 2 lists the same

attributes associated with longwall mines. Specific factors that

quantify these mine attributes during 1995-98 include the

number of hours worked underground, number of mines,

frequency of roof and rib fall injuries, and associated incidence

rates.

MINING METHOD

According to figure 2, the roof fall injury incidence rate for

room-and-pillar mines is more than double the longwall roof

fall rate. On the other hand, the rib incidence rate for the two

methods is nearly identical. The significantly lower longwall

roof fall rate may be related to the continuous roof protection

provided by the longwall face supports at the active mining face

and the greater number of support workers located outby the

mining face. The nearly identical rib fall rate probably reflects

the thicker seams that are mined using longwalls. Table 3

compares the two mining methods. It is interesting to note that

even though longwall mines represent only 6% of the

underground coal mines, they are significantly larger on

average, accounting for 48% of all the underground coal

produced and 22% of all roof falls. Room-and-pillar operations

represent 94% of the mines and account for 52% of the tonnage

and 78% of all roof falls. It is also of interest that the

percentage of reportable noninjury roof falls is very similar

Table 1.—Attributes of roof and rib fall injuries at room-and-pillar mines, 1995-98

Room-and-pillar characteristics

Hours worked Mines Roof falls2 Rib falls2 Incident rate

No. % No. % No. % No. % Roof falls per

200,000 hr

Rib falls per

200,000 hr

Location:

 Eastern PA . . . . . . . . . . . . . . . . . 1,090,084 0.5 180 5.0 2 0.1 3 0.9 0.37 0.55

 Western PA . . . . . . . . . . . . . . . . 16,332,658 7.6 189 5.3 134 6.1 15 4.4 1.64 0.18

 Northern WV/OH/MD . . . . . . . . . 9,829,896 4.6 185 5.2 85 3.8 9 2.6 1.73 0.18

 Central WV . . . . . . . . . . . . . . . . 51,605,701 24.2 908 25.5 504 22.8 114 33.4 1.95 0.44

 VA . . . . . . . . . . . . . . . . . . . . . . . 24,543,233 11.5 583 16.4 273 12.4 32 9.4 2.22 0.26

 Eastern KY . . . . . . . . . . . . . . . . 35,843,027 16.8 719 20.2 341 15.4 40 11.7 1.90 0.22

 Central KY and TN . . . . . . . . . . . 28,965,438 13.6 557 15.6 311 14.1 51 15.0 2.15 0.35

 IL/IN . . . . . . . . . . . . . . . . . . . . . . 22,411,104 10.5 76 2.1 314 14.2 36 10.6 2.80 0.32

 Western U.S. . . . . . . . . . . . . . . . 5,584,663 2.6 76 2.1 48 2.2 35 10.3 1.72 1.25

 Western KY . . . . . . . . . . . . . . . . 15,279,890 7.2 73 2.0 163 7.4 4 1.2 2.13 0.05

 AL . . . . . . . . . . . . . . . . . . . . . . . 2,150,106 1.0 19 0.5 35 1.6 2 0.6 3.26 0.19

Mine size:

 <50 workers . . . . . . . . . . . . . . . . 96,827,951 45.3 3,049 85.5 963 43.6 128 37.5 1.99 0.26

 50-149 workers . . . . . . . . . . . . . 80,891,498 37.9 439 12.3 855 38.7 145 42.5 2.11 0.36

 >149 workers . . . . . . . . . . . . . . . 35,916,351 16.8 77 2.2 392 17.7 68 19.9 2.18 0.38

Seam height:3

 <43 in . . . . . . . . . . . . . . . . . . . . . 60,797,201 28.5 1,581 44.3 540 24.4 56 16.4 1.78 0.18

 43-60 in . . . . . . . . . . . . . . . . . . . 86,382,352 40.4 1,177 33.0 957 43.3 115 33.7 2.22 0.27

 >60 in . . . . . . . . . . . . . . . . . . . . . 63,556,005 29.7 706 19.8 686 31.0 168 49.3 2.16 0.53

All room-and-pillar mines ....... 213,635,800 — 3,565 — 2,210 — 341 — 2.07 0.32 1

The total number of mines for 1995-98 is not mutually exclusive (e.g., if a mine operated all 4 years, it is counted four times). 2

All falls resulting in degree of injury of 1 to 6.

3

Approximately 1.3% of the hours worked were at mines that did not report or misreported the seam height and are excluded.
7

 Figure 2.—Roof and rib fall injury rates by mining method,

1995-98. (Source: MSHA data.)

Table 2.—Attributes of roof and rib fall injuries at longwall mines, 1995-98

Longwall characteristics

Hours worked Mines Roof falls2 Rib falls2 Incident rate

No. % No. % No. % No. % Roof falls per

200,000 hr

Rib falls per

200,000 hr

Location:

 Eastern PA . . . . . . . . . . . . . . . . . — — — — — — — — — —

 Western PA . . . . . . . . . . . . . . . . 23,198,139 16.1 29 11.7 114 17.8 29 13.5 0.98 0.25

 Northern WV/OH/MD . . . . . . . . . 35,971,363 24.9 57 23.1 100 15.6 20 9.3 0.56 0.11

 Central WV . . . . . . . . . . . . . . . . 14,040,850 9.7 23 9.3 80 12.5 18 8.4 1.14 0.26

 VA . . . . . . . . . . . . . . . . . . . . . . . 7,454,159 5.2 16 6.5 23 3.6 3 1.4 0.62 0.08

 Eastern KY . . . . . . . . . . . . . . . . 341,450 0.2 1 0.4 2 0.3 — — 1.17 —

 Central KY and TN . . . . . . . . . . . 3,537,063 2.5 7 2.8 24 3.8 8 3.7 1.36 0.45

 IL/IN . . . . . . . . . . . . . . . . . . . . . . 10,787,674 7.5 18 7.3 91 14.2 32 14.9 1.69 0.59

 Western U.S. . . . . . . . . . . . . . . . 16,881,139 11.7 56 22.7 35 5.5 62 28.8 0.41 0.73

 Western KY . . . . . . . . . . . . . . . . 5,886,406 4.1 9 3.6 75 11.7 10 4.7 2.55 0.34

 AL . . . . . . . . . . . . . . . . . . . . . . . 26,087,525 18.1 31 12.6 96 15.0 33 15.3 0.74 0.25

Mine size:

 <50 workers . . . . . . . . . . . . . . . . 376,146 0.3 9 3.6 3 0.5 — — 1.60 —

 50-149 workers . . . . . . . . . . . . . 10,374,482 7.2 46 18.6 40 6.3 29 13.5 0.77 0.56

 >149 workers ............... 133,435,140 92.5 192 77.7 597 93.3 186 86.5 0.89 0.28

Seam height:3

 <43 in . . . . . . . . . . . . . . . . . . . . . — — — — — — — — — —

 43-60 in . . . . . . . . . . . . . . . . . . . 34,490,334 23.9 55 22.3 145 22.7 33 15.3 0.84 0.19

 >60 in . . . . . . . . . . . . . . . . . . . . . 108,855,545 75.5 188 76.1 491 76.7 182 84.7 0.90 0.33

 All longwall ............. 144,185,768 — 247 — 640 — 215 — 0.89 0.30 1

The total number of mines for 1995-98 is not mutually exclusive (e.g., if a mine operated all 4 years, it is counted four times). 2

All falls resulting in degree of injury of 1 to 6.

3

Approximately 0.5% of the hours worked were at mines that did not report or misreported the seam height and are excluded.

to the percentage of roof fall injuries for both mining methods.

This may indicate that the geotechnical conditions that result in

roof falls may be a function of the mining method used. Since

there is such a large and distinct difference between the two

mining methods, the remaining characteristic evaluations will

break out the mining methods separately.

LOCATIONS

Examination of roof fall injury rates at longwall operations

(figure 3) in all coal mining districts in the United States found

that western Kentucky and Illinois/Indiana have roof fall injury

rates that are 3 and 1.7 times the national longwall average,

respectively. Conversely, the western United States, northern

West Virginia/Ohio/Maryland, and Virginia have a longwall

rate that is at least 25% less than the national average. To

determine if these same trends are observable with massive roof

falls, noninjury reportable roof falls were evaluated. Figure 4

shows a similar trend; western Kentucky and the Illinois Basin

were found to have significantly higher longwall noninjury roof

fall incident rates, and Alabama, Virginia, and northern West

Virginia/Ohio/Maryland had lower rates. High fatality rates

were identified in longwalls in southeastern Kentucky/

Tennessee, the western United States, and Illinois/Indiana,

while Virginia and western Kentucky had zero fatalities in

longwall mines (figure 5). Longwall rib fall rates were high in

the western United States, Illinois/Indiana, and southeastern

Kentucky/Tennessee and low in Virginia and northern West

Virginia/Ohio/Maryland (figure 6).

Locality trends of room-and-pillar operations found

Alabama and Illinois/Indiana with high roof fall injury rates

(figure 3); no location was found to a have significantly low

rates except eastern Pennsylvania. Noninjury roof fall rates

revealed that western Kentucky and northern West Virginia/

Ohio/Maryland exceeded the national room-and-pillar mine

average by 140% and 40% (figure 4), respectively, and eastern
8

Figure 3.—Roof fall injury rates by location, 1995-98. (Source: MSHA data.)

Table 3.—Comparison of mining methods, 1995-98

Attribute

Mining method

Room-and-pillar Longwall Combined

No. % No. %

Production:

 Tons produced ................. 861,172,448 52 783,644,012 48 1,644,816,460

 Underground hours worked ....... 213,635,800 60 144,185,768 40 357,821,568

 Workers . . . . . . . . . . . . . . . . . . . . . . 109,425 62 65,976 38 175,401

 Mines1 . . . . . . . . . . . . . . . . . . . . . . . 3,565 94 247 6 3,812

Groundfalls:

 Roof fall injuries2 . . . . . . . . . . . . . . . 2,210 78 640 22 2,850

 Rib fall injuries2 . . . . . . . . . . . . . . . . 341 61 215 39 556

 Roof fall noninjuries . . . . . . . . . . . . . 6,093 80 1,543 20 7,636

Incident rates:3

 Roof fall injuries per 200,000 hr . . . . 2.07 — 0.89 — 1.59

 Rib fall injuries per 200,000 hr . . . . . 0.32 — 0.3 — 0.31

 Roof fall noninjuries per 200,000 hr . 5.70 — 2.14 — 4.27

 1

The total number of mines for 1995-98 is not mutually exclusive (e.g., if a mine worked all 4 years,

it is counted four times).

 2

All falls resulting in degree of injury of 1 to 6.

 3

The combined rate is the total number of incidents divided by the total number of hours worked.
9

Figure 4.—Roof fall noninjury rates by location, 1995-98. (Source: MSHA data.)

Figure 5.—Roof and rib fall fatality rates by location, 1995-98. (Source: MSHA data.)
10

Figure 6.—Rib fall injury rates by location, 1995-98. (Source: MSHA data.)

Kentucky, eastern Pennsylvania, and Alabama were below the

national average. Figure 5 displays the high groundfall fatality

rates occurring at room-and-pillar mines in eastern Pennsylvania, Virginia, and the western United States and zero fatalities

in northern West Virginia/Ohio/Maryland, Alabama, and

Illinois/Indiana. The rib fall injury rate at room-and-pillar

mines was four times the national average for the western

United States; other high rates were found in eastern

Pennsylvania and central West Virginia. Low rib fall rates

occurred in western Kentucky, western Pennsylvania, and

northern West Virginia/Ohio/Maryland (figure 6). High

groundfall fatality and rib fall rates were found in eastern

Pennsylvania and may be associated with the unusual folded

stratigraphy of anthracite mines (figures 5-6). It is interesting

to note that room-and-pillar mines in northern West Virginia/

Ohio/Maryland had a noninjury reportable roof fall rate that

 is 40% higher than the national rate, yet the injury roof fall rate

is 16% lower than the average, its rib injury rate is 44% less

than average, and its fatality rate during this period was zero

(figures 3-6).

Table 4 gives an overview of all of the high (at least 25%

higher than the national average) and low (at least 25% lower

than the national average) rates, ranking the locality for each

type of groundfall rate. It is evident that for longwall mining,

Table 4.—Overview of coal mining locations with extreme groundfall rates, 1995-98

Mine type/rate Above the national rate by 25% Below the national rate by 25%

Room-and-pillar mines:

 Roof fall injury rate . . . . . . IL/IN, AL . . . . . . . . . . . . . . . . . . . . . . . . . . . . Eastern PA.

 Roof fall noninjury rate . . . Western KY, northern WV/OH/MD . . . . . . . . Eastern KY, AL, eastern PA.

 Roof/rib fatality rate . . . . . Eastern PA, VA, Western U.S. . . . . . . . . . . . IL/IN, northern WV/OH/MD, AL.

 Rib fall injury rate . . . . . . . Western U.S., eastern PA. central WV . . . . . Western KY, western PA, northern WV/OH/MD.

Longwall mines:

 Roof fall injury rate . . . . . . Western KY, IL/IN, southeastern KY/TN . . . . Western U.S., northern WV/OH/MD, VA.

 Roof fall noninjury rate . . . Western KY, IL/IN . . . . . . . . . . . . . . . . . . . . . Eastern KY, AL, VA, northern WV/OH/MD.

 Roof/rib fatality rate . . . . . Southeastern KY/TN, western U.S., IL/IN . . . VA, eastern KY, western KY.

 Rib fall injury rate . . . . . . . Western U.S., IL/IN, southeastern KY/TN . . . VA, northern WV/OH/MD.
11

western Kentucky, southeastern Kentucky/Tennessee, and

Illinois/Indiana are fairly consistently listed for each rate type,

indicating that these regions have higher risk of groundfall

hazards than any other parts of the country. Localities that had

consistently lower risk of groundfall hazards for longwall

operations include Virginia and northern West Virginia/Ohio/

Maryland. Eastern Kentucky is listed with low rates; however,

longwall activity in this region is almost negligible and is

excluded. Room-and-pillar operations indicate a higher risk in

the western United States and eastern Pennsylvania and a lower

risk in northern West Virginia/Ohio/Maryland and Alabama.

Combining both mining methods shown in table 4 reveals

that Illinois/Indiana and western Kentucky are consistently

listed for both mining methods and for nearly all groundfall rate

types. The unique coalbed conditions of this area may be an

overriding factor producing this regional trend and will be

discussed further in the "Coalbed" section below. Also evident

in table 4 is the high rib fall injury rate in the western United

States for both mining methods. The high groundfall fatality

rate in the western United States may also be attributed to fatal

rib falls. These occurrences in the western United States may

be related to the higher and unstable ribs in the western United

States, as well as deeper overburdens, which are more prone to

bump and burst.

SEAM HEIGHT AND MINE SIZE

Previous studies [Fotta et al. 1997] have examined relationships between underground coal mine injury rates and mining

height and have emphasized the importance of controlling the

analysis for mining method and mine size. Mines operating in

thin seams (<43 in) tend to be smaller mines that exclusively

use the room-and-pillar extraction method. All longwall mines

operate in medium or thick seams (>43 in) and are predominantly large mines (>149 workers). To control for mining

method, all longwall mines are excluded from the study. It

should be noted that there is not a wide distribution of mine

sizes and seam heights for longwalls mines, so excluding

longwalls will not overlook any seam height or mine size

trends. Controlling the study for room-and-pillar mines, mine

size, and seam height produces the groundfall incidence rate

trends shown in figures 7 through 10. The roof fall injury

incidence rates shown in figure 7 do not show any significant

trends (±25% of the national average). However, the noninjury

roof fall incidence rates (figure 8) reveal that small mines (<50

workers) and large mines (>149 workers) in thick seams

(>61 in) have a significantly greater risk of massive roof falls.

Conversely, mines in thin seams with small- and medium-sized

workforces have a significantly lower noninjury roof fall rate.

This trend slightly deviates with the groundfall fatality

incidence rates shown in figure 9. Small mines in thin seams

have a groundfall fatality rate that exceeds the national average

by 44%, whereas small mines located in thick seams have a

groundfall fatality rate that is 53% lower than the national

average. With regard to rib fall injury incidence rates

(figure 10), small- and medium-sized mines in thick seams

exceed the national average by over 100% and 60%,

respectively. Conversely, small- and medium-sized mines in

thin seams have a significantly lower rib fall incidence rate

(42% and 36% lower than the national average, respectively).

A comparison of all groundfall incidence rate trends in the

table 5 generally shows a higher risk of groundfalls for thick

seams for most mine sizes. By contrast, small thin-seam mines

have an extraordinarily high fatality rate. A reverse trend

occurs for groundfall rates of lower risk, particularly for smalland medium-sized mines in thin seams and for the fatality rate

of small mines in thick-seam mines. Perhaps the lack of cabs

and canopies in small, low seams results in higher groundfall

fatality rates, especially when massive falls occur. Another

explanation may be that not all of the groundfall injuries that

occur in small mines are reported.

Parallel trends occur for rib fall incidence rates, with high rib

fall rates at small- and medium-sized, thick-seam mines and

low rates in similar-sized mines with thin seams. As the mining

height increases, a greater surface area of the rib is exposed and

at risk of becoming unstable or prone to collapse. Perhaps these

high ribs are adequately supported in the large-sized mines,

whereas small- and medium-sized mines may have a minimal

staff to maintain the unstable ribs, little capital to purchase and

install rib bolts, or they may be located in unusual coalbeds

with complex ground control problems.

COALBED

To determine if certain coalbeds are more susceptible to

groundfalls, the coalbeds where groundfalls occurred were

evaluated. Since this is not a defined parameter in the MSHA

database, a listing of all underground coal mines and associated

coalbed names was obtained from the U.S. Department of

Energy's Energy Information Agency and merged with the

MSHA address information associated with every underground

coal mine. All MSHA district offices were surveyed to find any

missing coalbed names. Coalbed names were not identified for

15% of the room-and-pillar mines, which accounts for 2.4% of

room-and-pillar mine hours worked; these were mostly small

mines. All longwall coalbeds were accounted for.

Since more than 122 coalbeds were identified that produced

coal during 1995-98, a process was developed to select the

more significant coalbeds. Removed from the analysis were all

coalbeds with fewer than 400,000 hr worked or with fewer than

4 incidents (for each type of incident: roof fall injury, rib fall

injury, or noninjury reportable roof fall). The more significant

coalbeds were determined by calculating the percentage that the

coalbed groundfall rate exceeded the national average rate. All

coalbeds that exceeded that national rate by at least 25% are

listed regionally for each mining method in tables 6-7. The
12

 Figure 7.—Roof fall injury rates by mine size and seam height for room-and-pillar mines, 1995-98.

(Source: MSHA data.)

 Figure 8.—Roof fall noninjury rates by mine size and seam height for room-and-pillar mines, 1995-

98. (Source: MSHA data.)
13

 Figure 9.—Roof and rib fall fatality rates by mine size and seam height for room-and-pillar mines, 1995-

98. (Source: MSHA data.)

 Figure 10.—Rib fall injury rates by mine size and seam height for room-and-pillar mines, 1995-98.

(Source: MSHA data.)
14

Table 5.—Overview of mine size and seam height factors

with extreme groundfall rates, 1995-98

Mine type/rate Above the national rate

by 25%

Below the national rate

by 25%

Small mines (<50 workers):

 Roof fall injury rate . . . . . . . . None . . . . . . . . . . . . . . . . . . None.

 Roof fall noninjury rate . . . . . . Thick seam . . . . . . . . . . . . . Thin seam.

 Roof/rib fatality rate . . . . . . . . Thin seam . . . . . . . . . . . . . . Thick seam.

 Rib fall injury rate . . . . . . . . . . Thick seam . . . . . . . . . . . . . Thin seam.

Medium mines (50-149 workers):

 Roof fall injury rate . . . . . . . . . None . . . . . . . . . . . . . . . . . . None.

 Roof fall noninjury rate . . . . . . None . . . . . . . . . . . . . . . . . . Thin seam.

 Roof/rib fatality rate . . . . . . . . Thick seam . . . . . . . . . . . . . Thin seam.

 Rib fall injury rate . . . . . . . . . . Thick seam . . . . . . . . . . . . . Thin seam.

Large mines (>149 workers):

 Roof fall injury rate . . . . . . . . . None . . . . . . . . . . . . . . . . . . None.

 Roof fall noninjury rate . . . . . . Thick seam . . . . . . . . . . . . . Medium seam.

 Roof/rib fatality rate . . . . . . . . Thick seam . . . . . . . . . . . . . Thin seam.

 Rib fall injury rate . . . . . . . . . . None . . . . . . . . . . . . . . . . . . Medium seam.

NOTE: A thin seam is #42 in; a medium seam is 43-60 in; a thick seam is $61 in. "None"

indicates that seam thickness and mine size did not result in significant change (±25%) in

incidence rate.

broad regions shown in the tables do not match the more

specific locations listed earlier in the paper. Many of these

coalbeds in specific locations were found to overlap, and

broader regions were found to be a better indicator of defining

the location of the coalbed.

The most startling indication from previewing the room-andpillar coalbed listings in table 6 is the severe roof fall rates in

Illinois Basin coalbeds that affect 95% of the coal mines in that

region. This same trend is found in the longwall coalbed listing

for the Illinois Basin, shown in table 7, which affects 100% of

the longwall mines in the region. The notorious coalbeds

include the Kentucky No. 13, Herrin/No. 6/Kentucky No. 11,

and Springfield No. 5/Kentucky No. 9. These coalbeds have

injury and noninjury roof fall rates that significantly exceed the

national average for both mining methods, as well as high rib

fall injury rates for longwall mines. These high groundfall

incident rates in the Illinois Basin seem to correlate with the

high rates indicated in a previous section of this paper that

examined various mining localities, most notably Illinois/

Indiana and western Kentucky. Obviously, some unique set of

geological circumstances or stress fields are producing this

regional concentration of groundfalls.

Other troubling trends from this analysis are the high injury

and noninjury roof fall rates associated with 64% of the roomand-pillar mines in the northern Appalachian coalbeds,

specifically the Sewickley, Redstone, Pittsburgh, Bakerstown,

and Upper and Lower Freeport Coalbeds (table 6). These high

roof fall injury rates are not carried over into the longwall mines

of this region, except for the Upper Freeport and Sewickley

Coalbeds. Possibly the geology in this region is more flexible

and conducive for abutment load transfers that typically occur

with longwall mining. However, the longwall mines in the

Sewickley and Upper Freeport Coalbeds of the northern

Appalachian region, which represent only 10% of the mines,

have very high noninjury reportable roof fall rates. Perhaps the

long-term standup time for these coalbeds is considerably

shorter due to the high content of degradable shales that

comprise the immediate roof.

Coalbeds that engender difficult longwall mining conditions

seem to be located in the regions of central Appalachia and the

western United States. Over 55% of the longwall mines in the

central Appalachian region are located in coalbeds that have

groundfall incident rates that exceed the national average by at

least 25%. By contrast, only 29% of the room-and-pillar mines

in this same region have rates that exceed the national average.

Specifically problematic are the Eagle and Hazard No. 4

Coalbeds, which significantly exceed the national longwall rate

for roof fall injuries and noninjury roof fall injury rates.

Although the number of room-and-pillar mines in the central

Appalachian region with problematic coalbeds do not represent

a majority of mines, there are many coalbeds with high roof fall

rates, including the Ben Creek/Blair, Upper Banner, Jawbone/

Iaeger, and Walnut Mountain. In the same region, severe roomand-pillar rib fall injuries that exceed the national rate by over

100% include the Pocahontas No. 12, Eagle, Peerless,

Powellton, and Amburgy/Low Splint Coalbeds.

Another problematic area is the western United States, with

high rib fall rates that occur in 80% of the region's thick-seam

longwall mines. Longwall coalbeds that have rib fall rates that

exceed the national rate by over 100% include the Upper

Hiawatha, Hiawatha, Blind Canyon, Wadge, and B Seam.

Room-and-pillar coalbeds with high rib fall rates include the

D Seam and B Seam. The thick seams and deep overburdens

associated with the western United States probably contribute

to the rib control problems and make the rib faces more prone

to mountain bumps.
15

Table 6.—Room-and-pillar coalbeds with extreme groundfall rates, 1995-98

Region/coalbed1

DOE-EIA

coalbed

ID No.

Underground,

hr

No. of

mines2

Regional

mines represented, %

Percentage above

 national rate

Roof

injury

Rib

injury

Roof

noninjury

Northern Appalachian:

 Sewickley . . . . . . . . . . . . . . . . . 29 727,722 26 — — — 223

 Redstone . . . . . . . . . . . . . . . . . 33 830,127 23 — 51 — 82

 Pittsburgh . . . . . . . . . . . . . . . . . 36 1,960,405 37 — 78 — 57

 Bakerstown/Freeport . . . . . . . . 62 401,283 6 — 45 — —

 Upper Freeport . . . . . . . . . . . . . 71 7,783,479 82 — — — 25

 Lower Freeport . . . . . . . . . . . . . 74 1,329,040 14 — — — 45

 Upper Kittanning . . . . . . . . . . . 76 6,505,446 53 — — — 27

 Total . . . . . . . . . . . . . . . . . . . — 19,537,502 241 64.44 — — —

Central Appalachian:

 Hazard No. 8 . . . . . . . . . . . . . . 100 677,371 11 — 28 — —

 Hazard No. 7/High Splint . . . . . 104 592,004 14 — — — 54

 Coalburg/Hazard No. 6 . . . . . . . 111 11,558,414 143 — 30 — —

 Winifrede/Hazard No. 5 . . . . . . 121 4,138,977 45 — — 51 —

 Hatfield/No. 9 . . . . . . . . . . . . . . 127 614,743 3 — — — 25

 Walnut Mountain . . . . . . . . . . . 128 488,422 12 — — — 180

 Hernshaw/Whitesburg . . . . . . . 137 1,787,018 37 — 41 — —

 Amburgy/Low Splint . . . . . . . . . 142 6,700,453 112 — — 96 —

 Peerless . . . . . . . . . . . . . . . . . . 167 1,352,385 18 — — 132 —

 Powellton . . . . . . . . . . . . . . . . . 170 4,253,833 34 — — 121 —

 Eagle . . . . . . . . . . . . . . . . . . . . 176 2,614,897 48 — — 164 33

 Bens Creek/Blair . . . . . . . . . . . 177 890,525 17 — 139 — 136

 Glamorgan . . . . . . . . . . . . . . . . 185 1,855,293 52 — 35 — —

 Splash Dam . . . . . . . . . . . . . . . 210 3,504,704 85 — — 25 —

 Upper Banner . . . . . . . . . . . . . . 214 904,347 36 — 60 — 28

 Jawbone/Iaeger . . . . . . . . . . . . 266 3,755,667 84 — 26 — 38

 Lower Iaeger/No. 4 . . . . . . . . . . 269 476,014 15 — — — 47

 Pocahontas No. 12 . . . . . . . . . . 311 2,939,232 27 — — 199 —

 Total . . . . . . . . . . . . . . . . . . . — 49,104,299 793 28.66 — — —

Southern Appalachian:

 Gholson . . . . . . . . . . . . . . . . . . 223 1,182,661 4 — 137 — —

 Total . . . . . . . . . . . . . . . . . . . — 1,182,661 4 21.05 — — —

Illinois Basin:

 Danville/No. 7 . . . . . . . . . . . . . . 480 2,228,338 8 — 47 — —

 KY No. 13 . . . . . . . . . . . . . . . . 482 784,239 4 — 147 — 937

 Herrin/No. 6/KY No. 11 . . . . . . 484 14,943,246 58 — 36 — —

 No. 5/Springfield/KY No. 9 . . . . 489 18,216,127 63 — — — 87

 Western KY No. 4 . . . . . . . . . . 520 900,097 9 — — — 99

 Total . . . . . . . . . . . . . . . . . . . — 37,072,047 142 95.30 — — —

Western United States:

 B Seam . . . . . . . . . . . . . . . . . . 1753 744,443 5 — — 574 —

 D Seam . . . . . . . . . . . . . . . . . . 1755 621,980 6 — 102 1,210 47

 Cameo . . . . . . . . . . . . . . . . . . . 1770 401,341 7 — — — 127

 Lower O'Connor ............ 1830 675,148 4 — 29 — 56

 Total . . . . . . . . . . . . . . . . . . . — 2,442,912 22 28.95 — — — 1

This analysis excludes coalbeds with fewer than 400,000 hr worked and fewer than four groundfall incidents.

2

The total number of mines for 1995-98 is not mutually exclusive (e.g., if a mine worked all 4 years, it is counted four times).
16

Table 7.—Longwall coalbeds with extreme groundfall rates, 1995-98

Region/coalbed1

DOE-EIA

coalbed

ID No.

Underground,

hr

No. of

mines2

Regional

mines represented, %

Percentage above 

national rate

Roof

injury

Rib

injury

Roof

noninjury

Northern Appalachian:

 Sewickley . . . . . . . . . . . . . . . . . . 29 741,827 2 — — — 505

 Upper Freeport . . . . . . . . . . . . . . 71 2,737,966 7 — 73 — 200

 Total . . . . . . . . . . . . . . . . . . . . — 3,479,793 9 10.47 — — —

Central Appalachian:

 Hazard No. 4 . . . . . . . . . . . . . . . 135 1,729,817 4 — 95 — 30

 Alma/Elkhorn No. 1/Blue Gem . . 157 2,193,175 6 — 34 84 —

 Imoboden/Warfield . . . . . . . . . . . 168 1,259,612 5 — 43 — —

 Eagle . . . . . . . . . . . . . . . . . . . . . 176 5,093,936 11 — 46 71 71

 Total . . . . . . . . . . . . . . . . . . . . — 10,276,540 26 55.32 — — —

Southern Appalachian:

 Pratt/Corona . . . . . . . . . . . . . . . . 227 2,040,393 4 — — — 143

 Total . . . . . . . . . . . . . . . . . . . . — 2,040,393 4 12.90 — — —

Illinois Basin:

 KY No. 13 . . . . . . . . . . . . . . . . . 482 3,254,456 4 — 357 44 285

 Herrin/No. 6/KY No. 11 . . . . . . . 484 6,780,473 14 — 49 88 148

 Springfield/No. 5/KY No. 9 . . . . . 489 6,639,151 9 — 87 62 144

 Total . . . . . . . . . . . . . . . . . . . . — 16,674,080 27 100.00 — — —

Western United States:

 Wattis . . . . . . . . . . . . . . . . . . . . . 1236 1,225,605 3 — — — 83

 Wadge/Roland of Tuff ........ 1750 2,388,299 4 — — 153 —

 B Seam . . . . . . . . . . . . . . . . . . . 1753 2,692,812 9 — — 149 87

 Lower O'Connor ............. 1830 2,264,391 8 — — 78 —

 Castle Gate B/Upper O'Connor . 1832 829,604 4 — — — 46

 Hiawatha . . . . . . . . . . . . . . . . . . 1846 2,340,148 9 — — 187 —

 Upper Hiawatha ............. 1847 1,059,750 4 — — 280 —

 Blind Canyon . . . . . . . . . . . . . . . 1855 1,694,672 4 — — 256 —

 Total . . . . . . . . . . . . . . . . . . . . — 14,495,281 45 80.36 — — — 1

This analysis excludes coalbeds with fewer than 400,000 hr worked and fewer than four groundfall incidents.

2

The total number of mines for 1995-98 is not mutually exclusive (e.g., if a mine worked all 4 years, it is counted four times).

SEASONAL PATTERNS

The chronological quarterly groundfall rates were evaluated

to determine if seasonal patterns, such as fluctuations in

temperature, barometric pressure, and humidity, might affect

the number of groundfall incidents. Since the western United

States has mostly an arid climate with minimum fluctuations in

humidity, incidents in Colorado, New Mexico, Utah and

Wyoming were excluded. Although monthly production data

are not compiled, quarterly hours worked were accessed from

MSHA's Terra database. This allowed the groundfall incidents

to be normalized based on the quarterly employee hours worked

underground. According to figure 11, the roof fall injury

incidence rate is fairly consistent, except for the third quarter

(July to September), where the incident rate peaks 30% higher

than the other three quarters. A similar pattern occurs with the

noninjury reportable roof fall rate, as shown in figure 12. The

fall rate is fairly consistent until the third quarter, where the

noninjury roof fall incident rate peaks 48% higher than the first

two quarters. Using the noninjury reportable roof fall rate is an

even better indicator of unstable ground conditions, since they

usually result in massive falls and are required to be reported.

Possibly this trend shows that mine air becomes more humid

during the summer months and the moisture is disintegrating

the shale roof, resulting in large groundfalls. It is interesting to

note that in figure 12 the noninjury fall rate for the fourth

quarter (October to December) is slightly elevated compared to

the first two quarters. Perhaps the third quarter trend is

continuing into the first month of the fourth quarter (October).

Other studies have found similar seasonal patterns.

Stateham and Radcliffe [1978] found that humidity has a strong

influence on roof fall occurrence rates. Their results indicated

that the probability of a roof fall is highest in August and lowest

in February.
17

 Figure 11.—Roof and rib fall injury rates by quarter, excluding the western United States, 1995-98.

(Source: MSHA data.)

 Figure 12.—Roof fall noninjury rates by quarter, excluding the western United States, 1995-98.

(Source: MSHA data.)
18

OVERVIEW OF FATALITY REPORTS ASSOCIATED WITH GROUND CONTROL HAZARDS

During 1996-99, 49 underground coal miners were killed in

46 separate incidents. Table 8 lists the frequency of primary

and secondary factors that contributed to these incidents. In

some cases, more than one hazard was involved. For example,

12 fatalities occurred during pillar extraction; three of the

incidents resulted from premature collapses in intersections.

UNSUPPORTED ROOF

When miners go under unsupported roof, they are completely unprotected. According to table 8, approximately 24%

of all roof fall fatalities during the study period occurred when

miners traveled under unsupported roof. While there are no

grounds for complacency, the recent record does represent an

improvement from a decade ago, when nearly 50% of groundfall fatalities occurred beneath unsupported roof [Peters 1992].

The improvement was achieved through new equipment,

enforcement, and a persistent educational campaign.

By definition, roof support activities occur very close to

unsupported roof. Therefore, it is not surprising that most of

the fatal accidents involved roof bolt operators or other miners

engaged in roof support. Based on the accident record, singlehead roof bolt machines seem to be a risk factor. Roof control

plans carefully specify the sequence of bolt installation with

single-head machines to avoid placing the operator inby

support. If these guidelines are not followed, the roof bolt

operator can be at risk.

During the early 1990s, the USBM conducted an extensive

series of interviews with miners to determine why they risk

going under unsupported roof [Peters 1992]. The most common

response was that they had unintentionally walked out beyond

the supports. The most effective countermeasure, then, is to

ensure that all areas of unsupported roof are clearly posted with

highly visible warning devices.

Relatively simple procedures or technologies can be

implemented to reduce the temptation for workers to intentionally go beyond support. However, training is also

essential. Other USBM studies [Mallett et al. 1992] argue that

verbal admonitions and threats of discipline are less effective

than training that graphically imparts the severe consequences

of roof falls. A series of three videos was prepared in which

actual miners are interviewed about roof fall accidents that they

had experienced. The videos also emphasize the impact of roof

fall accidents on people other than the one caught in the fall.

These highly effective videos, together with training manuals,

are available from MSHA's National Mine Health and Safety

Academy near Beckley, WV.

ROOF SKIN FAILURES

Skin failures are incidents that do not involve failure of the

roof support elements, but result from rock spalling from

between roof bolts, around ATRS systems, or from ribs. They

are of particular concern because they cause injuries and

fatalities to workers who should have been protected by

supports. In 1997, 98% of the 810 roof and rib injuries suffered

by mine workers were attributed to skin failures [Bauer et al.

1999]. Because groundfall fatalities are usually the result of

massive roof failure, this study found only 12% of the roof falls

were related to smaller scale roof skin failures.

Roof skin failures almost always involve pieces of rock that

are less than 2 ft thick. About 40% of the 669 roof skin injuries

in 1997 involved roof bolt operators and occurred beneath

temporary support. The other roof skin injuries occurred

beneath permanent support and involved workers in a wide

variety of activities. Common roof skin control techniques

include oversized plates, header boards, wood planks,

steel straps, mesh, and (in rare instances) spray coatings

(sealants).

RIB FAILURES

During 1996-99, rib failures resulted in seven fatalities, or

14% of all groundfall fatalities, as shown in table 8. Only one

Table 8.—Hazards associated with fatal groundfalls, 1996-99

Hazard No. of fatalities Percentage

Primary Secondary1 Total Primary Secondary1 Total

Geologic . . . . . . . . . . . . . . 1 10 11 2 20 22

Roof skin . . . . . . . . . . . . . . 6 0 6 12 0 12

Rib . . . . . . . . . . . . . . . . . . 7 2 9 14 4 18

Pillaring . . . . . . . . . . . . . . . 12 2 14 24 4 29

Inby unsupported roof . . . . 12 0 12 24 0 24

Intersection . . . . . . . . . . . . 3 5 8 6 10 16

Longwall face .......... 4 0 4 8 0 8

Construction . . . . . . . . . . . 4 3 7 8 6 14

 Total . . . . . . . . . . . . . . . 49 22 71 100 44 —

1

A secondary hazard was assigned to some groundfall fatalities, but not in all cases.
19

of these fatal injuries was to a face worker, while five were

mechanics and electricians performing their duties well outby

the face. Nearly 80% of the 128 rib injuries that occurred in

1997 took place beneath permanently supported roof. Nonfatal

rib injuries resulted in an average of 43 lost workdays each

versus 25 days for the average roof skin injury.

Seam height is the single greatest factor contributing to rib

failures. The seam height was >8 ft in all six of the fatalities

and >10 ft in three of them. The incidence of rib injuries

increases dramatically once the seam height reaches 7 ft. No

rib support was used in any of the six fatal accidents.

Rib failure is often associated with rock partings and/or

discontinuities within the pillar or with overhanging brows

created by roof drawrock. The most effective rib supports use

full planks or mesh held in place by roof bolts.

PILLAR RECOVERY

The process of pillar recovery removes the main support for

the overburden and allows the ground to cave. As a result, the

pillar line is an extremely dynamic and highly stressed

environment. Safety depends on controlling the caving through

proper extraction sequencing and roof support. In some mines,

mobile roof supports have replaced timber supports for each

stage of pillar recovery.

According to table 8, pillaring fatalities are directly

attributed to 24% of all groundfall fatalities, including three

multiple incidents. However, a recent study estimated that

pillar recovery accounts for only 10% of the coal mined

underground [Mark et al. 1997]. Nearly 50% of these fatal

pillaring incidents involved geologic discontinuities, such as

slips and slickensides. Even mines that used additional support

for these discontinuities were unable to prevent the massive

roof failures. These failures often occur suddenly with little

warning and result in collapses where MSHA is unable to find

any violations. During 1987-96, Mark et al. [1997] found that

almost 50% of the pillaring fatalities occurred during the

recovery of the final lift or pushout. Since 1996, however, only

20%, or three incidents, involved last-lift incidents. Several

incidents also involved situations where excessive cuts were

taken, which caused the large exposed intersection to collapse.

Pillar recovery can also be difficult under deep cover.

During 1996-98, nearly one-half of the pillar recovery fatalities

occurred where the depth of cover exceeded 650 ft. Under deep

cover, barrier pillars and special mining sequences may be

required.

GEOLOGIC DISCONTINUITIES

Geologic features such as slips, slickensides, clay veins,

kettlebottoms, and ancient stream channels have been closely

linked with many groundfall fatalities. These hazardous

geologic structures are found predominantly in the Appalachian

coal basins [Chase 1992]. Geologic discontinuities were mostly

identified as the secondary contributing causes of failure in 20%

of all groundfall fatalities (table 8).

Slickensides and slips were found to be the primary and

secondary causes of failure of four similar massive groundfalls

in 1997 that resulted in four fatalities and seven injuries. All of

the falls were so massive that they overran the permanent

support, resulting in the collapse of the bolted intersection.

Two of the occurrences were attributed primarily to pillaring,

which triggered the unstable slips and slickensided joints to

collapse. Special precautions need to be taken near the outcrop,

where the presence of groundwater and weathered joints

(sometimes called hill seams) can reduce roof competence. In

general, pillar recovery should not be conducted when the

distance to the outcrop is <150 ft.

LONGWALL FACES

The longwall system of mining, which extracts immense coal

panels and allows the roof to cave behind the face, presents a

unique ground control situation. The total extraction of the coal

causes high stress concentrations along the face and in the gate

roads and may pose severe ground control problems depending

on the competency of the immediate and main roof rock and the

sizing of the gate road pillars [Listak and Pappas 1990].

According to figure 8, approximately 8% of groundfall

deaths are associated with longwall face mining. Two incidents

involved a similar work activity of installing wire mesh in

preparation for recovering the longwall face equipment.

Usually, a large redistribution of stresses occurs as the longwall

face approaches the recovery room, which may weaken the roof

directly above the face. In the first incident, a slickensided

piece of top coal fell from the face and struck the victim located

under the shield supports. In the second case, a piece of binder

rock fell from the face, hitting the victim located between the

pan line and the longwall face.

CONSTRUCTION

Construction relates to any type of outby mining or

resupport of the coal or roof strata. Examples include cutting

the roof higher to install an overcast or belt line (boom hole) or

rehabilitating roof fall areas.

It seems unlikely that these types of incidents would happen

with multiple frequency. However, during the study period

seven construction fatalities, or 14% of groundfall fatalities,

occurred as either a primary or secondary cause (table 8). Four

of the construction incidents were related to boom holes, two to

overcast construction, and one to rehabilitating a high roof fall

area. Several of the boom hole and overcast incidents occurred

because the ATRS was not of sufficient height to support the

roof. This resulted in the use of other temporary support methods where the procedure was not properly followed. Also, two

incidents occurred while the victim went under unsupported

roof following boom hole shots. There seems to be some
20

confusion concerning the proper procedure in supporting boom

holes. The rehabilitation incident occurred while installing steel

arches. The victim was under the last arch that had been

installed and slid a mud sill under the unsupported roof when

the roof fell. The fall struck the inby edge of the last arch, then

toppled under the arches.

INTERSECTION STABILITY

Thousands of intersections are driven each year and create

diagonal spans of 25-40 ft, well over the normal width of an

entry. The hazards of wide spans can increase when pillar

corners are rounded for machine travel (turnouts) or when rib

spalling increases the span. According to table 8, 6% of roof

fall fatalities are primarily caused by oversized intersections and

10% are a secondary cause. In 1996, there were 2,105 noninjury reportable roof falls. More than 71% of these occurred

in intersections despite the fact that intersections probably

account for less than 25% of all drivage underground.

Intersection spans are often measured as the sum of the

diagonals. Because the rock load increases in proportion to the

cube of the span, even a small increase in the span can greatly

reduce the stability of an intersection. For example, widening

the entry from 18 to 20 ft increases the rock load from 96 to 132

tons. A study at a mine in western Pennsylvania found that

83% of the roof falls occurred in 13% of the intersections where

the sum of the diagonals exceeded 70 ft [Molinda et al. 1998].

Many roof control plans specify the maximum spans that are

allowed. Mining sequences can also be designed to limit the

number, location, and size of turnouts and to restrict turnouts to

specific entries. Extra primary support, such as longer roof

bolts, installed within intersections can also be very effective in

reducing the likelihood of roof falls. On the other hand,

replacing four-way intersections with three-ways may be not be

an effective control technique. Three-way intersections are

more stable, but since it normally takes two three-way

intersections to replace one four-way intersections, the total

number of falls is likely to increase [Molinda et al. 1998].

CONCLUSIONS

The effects of groundfall incidents are extensive, ranging

from the economic loss of equipment and production to fatal

and nonfatal injuries that result in lasting physical and financial

impairments suffered by the victims and the victim's family. In

addition, the mining industry is severely impacted by these

injuries, as well as thousands of noninjuries that damage

equipment, stop production, or disrupt ventilation. This study

of roof and rib fall injuries and noninjury rates controlled for

mining method, seam height, and mine size, and resulted in the

identification of the following incident trends:

$ The longwall mining method results in less than one-half

the roof fall injury rate compared to than the room-and-pillar

method. However, the rib injury rate for both mining methods

is nearly identical.

$ Longwall mining accounts for 48% of the production and

40% of the hours worked, but results in only 22% of the roof

fall injuries.

$ Longwall mines in western Kentucky and Illinois/Indiana

have significantly higher roof fall rates. Northern West

Virginia/Ohio/Maryland, Virginia, and the western United

States have significantly lower roof fall rates.

$ Room-and-pillar mines in western Kentucky have a very

high noninjury roof fall rate.

$ For both mining methods, rib fall injury rates are

significantly higher in the western United States.

$ For noninjury falls in room-and-pillar mines, small mines

(<50 workers) and large mines (>149 workers) in thick seams

(>60 in) have a significantly higher risk of massive roof falls.

Conversely, mines located in thin seams with small- and

medium-sized workforces have a significantly lower massive

fall rate.

$ The fatality rate for room-and-pillar mines is very high for

small mines in thin seams, but is very low for small mines in

thick seams.

$ Room-and-pillar mines in small- and medium-sized mines

in thin seams have a significantly lower rib fall rate. Small- and

medium-sized mines in thick seams have a significantly higher

rate.

$ For coalbeds in which both methods are used, severe

groundfall rates were identified in the Illinois Basin, especially

for the Kentucky No. 13, Herrin/No. 6/Kentucky No. 11, and

Springfield No. 5/Kentucky No. 9 Coalbeds.

$ For room-and-pillar coalbeds, northern Appalachian

coalbeds, most notably the Sewickley, Redstone, Pittsburgh,

and Bakerstown, have severe roof fall rates. Many coalbeds in

the central Appalachian region have high roof fall rates,

especially Bens Creek/Blair, Upper Banner, Jawbone/Iaeger,

and Walnut Mountain.

$ Longwall coalbeds in the central Appalachian coalfields

with very high groundfall rates include the Eagle and Hazard

No. 4.

$ Severe rib fall rates were found for several coalbeds in the

western United States. Difficult room-and-pillar coalbeds

include the D Seam and B Seam. Problematic longwall

coalbeds include the Upper Hiawatha, Hiawatha, Blind Canyon,

Wadge and B Seam.
21

$ Severe room-and-pillar rib fall rates were found in

coalbeds in the central Appalachian region, including the

Pocahontas No. 12, Eagle, Peerless, Powellton, and Amburgy/

Low Splint Coalbeds.

$ A review of seasonal patterns revealed that the third

quarter (July to September) has a 30%-40% increased risk of

injury and noninjury groundfalls. This may be due to higher

humidity levels.

To better understand why these groundfalls occurred, all of

the detailed fatality reports were analyzed, and the following

groundfall hazards contributing to the groundfall were

identified:

$ Pillaring is the leading cause of fatal groundfall failures.

Many of these incidents were triggered by geologic discontinuities present in the roof strata, such as slips and

slickensides. These failures often occurred suddenly and with

little warning.

$ A prevailing factor contributing to groundfall fatalities

was miners going under unsupported roof. In recent years,

fatalities from this risky activity has dropped significantly;

however, it still contributes to 24% of all groundfall fatalities.

$ Rib failure is a major hazard associated with groundfall

fatalities and is often associated with rock partings and/or

discontinuities within the pillar or with overhanging brows

created by roof drawrock. None of the mines where rib

fatalities occurred used any type of rib mesh or bolting.

$ Roof skin failures are of particular concern because they

caused 12% of groundfall fatalities and many nonfatal injuries

to workers who should have been protected by supports.

$ Construction-related groundfalls were associated with

14% of the primary or secondary causes of the fatalities.

Several boom hole and overcast incidents occurred because the

ATRS was not of sufficient height to support the roof.

$ Several longwall-mining-related groundfall fatalities

resulted during installation of wire mesh as the longwall face

approached the recovery room.

These groundfall statistical characteristics and fatality report

trends offer the most current profile of roof and rib falls in the

United States. This study identifies areas where additional

research is needed so that innovative solutions can be developed

to reduce these severe hazards to underground coal mine

workers.

REFERENCES

Bauer ER, Pappas DM, Dolinar DR, McCall FE, Babich DR [1999]. Skin

failure of roof and rib in underground coal mines. In: Peng SS, Mark C, eds.

Proceedings of the 18th International Conference on Ground Control in

Mining. Morgantown, WV: West Virginia University, pp. 108-114.

Chase FE [1992]. Geologic structures that affect Appalachian coal mines.

In: Proceedings of the Preventing Coal Mine Groundfall Accidents; How

To Identify and Respond to Geologic Hazards and Prevent Unsafe Worker

Behavior. Pittsburgh, PA: U.S. Department of the Interior, Bureau of Mines,

IC 9332, pp. 3-14.

Fotta B, Mallett LG [1997]. Effect of mining height on injury rates in U.S.

underground nonlongwall bituminous coal mines. Pittsburgh, PA: U.S.

Department of the Interior, Bureau of Mines, IC 9447.

Fotta B, Turin FC, Murphy JN [1995]. Appalachian coal-mining health

and safety trends. Min Eng 47(12):1115-1119.

Listak JM, Pappas DM [1990]. Longwall automation: a ground control

perspective. Pittsburgh, PA: U.S. Department of the Interior, Bureau of Mines,

IC 9244.

Mallett LG, Vaught C, Peters RH [1992]. Training that encourages miners

to avoid unsupported roof. In: Preventing Coal Mine Groundfall Accidents;

How To Identify and Respond to Geologic Hazards and Prevent Unsafe

Worker Behavior. Pittsburgh, PA: U.S. Department of the Interior, Bureau of

Mines, IC 9332, pp. 32-45.

Mark C, McCall FE, Pappas DM [1997]. A statistical overview of retreat

mining of coal pillars in the U.S. In: Peng SS, ed. Proceedings of the 16th

International Conference on Ground Control in Mining. Morgantown, WV:

West Virginia University, pp. 204-210.

Molinda GC, Bauer ER, Babich DR, Pappas DM [1998]. Factors

influencing intersection stability in U.S. coal mines. In: Peng SS, ed.

Proceedings of the 17th International Conference on Ground Control in

Mining. Morgantown, WV: West Virginia University, pp. 267-275.

Pappas DM [1987]. Roof and rib fall accident and cost statistics: an indepth study. Pittsburgh, PA: U.S. Department of the Interior, Bureau of Mines,

IC 9151.

Peters RH [1992]. Miners views on how to prevent people from going

under unsupported roof. In: Preventing Coal Mine Groundfall Accidents; How

To Identify and Respond to Geologic Hazards and Prevent Unsafe Worker

Behavior. Pittsburgh, PA: U.S. Department of the Interior, Bureau of Mines,

RI 9332, pp. 25-31.

Peters RH, Randolph RF [1991]. Miners' views about why people go under

unsupported roof and how to stop them. Pittsburgh, PA: U.S. Department of

the Interior, Bureau of Mines, IC 9300.

Stateham RM, Radcliffe DE [1978]. A cyclic effect in coal mine roof

stability. Washington, DC: U.S. Department of the Interior, Bureau of Mines,

RI 8291.
23

FUNDAMENTALS OF COAL MINE ROOF SUPPORT

By Christopher Mark, Ph.D.,1 and Thomas M. Barczak2

ABSTRACT

Roof supports can only be understood in conjunction with the rock structure that they support. The strength

of the rock depends on geology, and the loads are applied primarily by the in situ and mining-induced stresses.

Other factors, such as wider spans and retreat or multiple-seam mining, can also reduce the stability of mine

openings. Roof supports are used to help stabilize these openings, but their performance characteristics must

be properly matched to the loading environment and ground behavior if they are to succeed. Roof supports

include both intrinsic supports, such as roof bolts, and standing supports. The key characteristics of any

support include its maximum load-carrying capacity, stiffness, and residual strength. Other important factors

are the timing of installation, the stability of the support as it is loaded, and the capability of the support system

to provide skin control. This paper explains in practical terms how supports work and the important factors

in ensuring that a good support design and application strategy are developed.

1

Supervisory physical scientist.

2

Research physicist.

Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
24

INTRODUCTION

Roof support is essential to the safety of every underground

miner. It has three primary functions:

$ To prevent major collapses of the mine roof;

$ To protect miners from small rock falls that can occur

from the immediate roof skin; and

$ To control deformations so that mine openings remain

serviceable for both access and escape, as well as for ventilation

of the mine workings.

Roof supports interact with the ground to create a stable rock

structure. With any structure, an engineering analysis begins

with evaluations of two fundamental factors:

$ The strength of the different components of the structure;

and

$ The forces that are loading it.

Rock structures are unique in that the strength of one essential

component, the rock itself, can seldom be determined accurately. Similarly, the ground stresses are rarely well understood.

Ground control engineers have had to develop novel techniques

to compensate for these deficiencies.

This paper begins with a summary of the factors affecting

the integrity of mine roof structures. Next, it discusses the

function and properties of mine roof support. It concludes with

a framework for understanding how the supports and the

ground interact with each other to provide a stable mine

opening.

FACTORS AFFECTING THE INTEGRITY OF MINE STRUCTURES

An assessment of the integrity of any mine structure must

begin with an analysis of (1) the structural integrity and strength

of the roof rock, (2) the excavation geometry, and (3) the forces

applied to the mine roof.

ROCK STRENGTH

Rock strength traditionally is estimated from laboratory

tests. The uniaxial compressive test is the most commonly

used. Figure 1 shows the approximate range of compressive

strengths observed in U.S. coal measure rock. Triaxial tests,

where the rock is confined, more accurately simulate the threedimensional stress that rock typically encounters underground.

Shear tests of bedding planes can be very helpful in evaluating

the likelihood of slip, but are rarely performed in the United

States. These three types of tests are shown in figure 2.

Rock tests are severely limited in that they are conducted on

small samples of intact rock. The strength of the rock mass in

mine roof is, however, determined largely by the presence of

cracks, bedding planes, and other natural discontinuities. Rock

mass classification systems were developed to help quantify their

effects.

The Coal Mine Roof Rating (CMRR) focuses on the specific

features that commonly occur in coal measure rock. It weighs

the individual geotechnical factors that determine roof

competence, includingC

$ The uniaxial compressive strength of the intact rock;

$ The spacing and persistence of discontinuities like bedding planes and slickensides;

$ The cohesion and roughness of the discontinuities; and

$ The presence of ground water and the moisture sensitivity

of the rock.

Simple index tests and observations are used to rate each of

these parameters, which are then combined into a single rating

on a scale from 0 to 100.

The CMRR can be calculated from underground exposures

like roof falls and overcasts [Molinda and Mark 1994] or from

exploratory drill core [Mark and Molinda 1996]. In the case of

drill core, point load tests are used to estimate the compressive

strength and the cohesion. A computer program is currently being developed to aid in the collection, interpretation, and presentation of CMRR data.

The CMRR incorporates most of the geologic factors that

affect the mine roof. It does not address large-scale features,

like faults, sandstone channel margins, or igneous dikes. Such

features may cause major disruptions in relatively small areas

and should be treated individually.

CMRR values have been obtained from hundreds of coal

mines throughout the United States and abroad. Figure 3 shows

that the northern Appalachian coalfields typically have the

weakest roof in the United States; the strongest roof is found in

Utah. Ground conditions and roof bolt densities from three major coal mining countries are compared in figure 4 [Mark

1999b]. Roof bolt design guidelines are presented elsewhere in

these Proceedings [Mark 2000].

ROOF SPAN

In underground coal mining, the excavation geometry does

not vary much, but the span can be very important. The basic

principle that governs the relationship between stability and the

span was first formulated by Austrian tunneling engineers

[Bieniawski 1989]:
25

Figure 1.—Range of compressive strength for U.S. coal measure rocks [Rusnak and Mark 2000].

Figure 2.—Three types of laboratory strength tests. A, uniaxial compressive strength test; B, triaxial compressive strength test;

C, bedding plane shear test.
26

Figure 3.—Range of Coal Mine Roof Ratings (CMRR) observed in the United States [Molinda and

Mark 1994].

Figure 4.—Roof bolt densities observed in three coal mining countries [Mark 1999b].
27

Figure 5.—Relationship between CMRR, depth of cover, and the stability of extended

cuts [Mark 1999a].

$ For a given rock mass, a tunnel's standup time decreases

as the roof span becomes wider; and

$ For a given roof span, a tunnel's standup time decreases

as the rock mass quality becomes poorer.

The greatest spans in coal mines are encountered in

intersections. While entries are normally limited to 6 m (20 ft),

the diagonal spans of intersections are generally in the 7.5-12

m (25-40 ft) range. Approximately 70% of all roof falls occur

in intersections, although intersections only account for about

20% to 25% of all drivage. Roof falls are therefore 8 to

10 times more likely to occur in intersections than in an

equivalent length of entry [Molinda et al. 1991].

A study by Mark [1999a] looked at standup time during extended (deep) cut mining, where the continuous miner advances

the face more than 20 ft beyond the last row of permanent

supports. At 36 mines, it was found that when the CMRR was

>55, the roof was stable in nearly every case. When the CMRR

was <37, the roof collapsed before the cut could be completed.

When the CMRR was between 38 and 55, extended cuts were

feasible some times, but not others (figure 5). The data also

show that extended cuts are less likely to be stable if either the

entry span or the depth of cover is increased.

Many studies have documented the effect of roof span on

stability. The longwall study cited earlier found a strong

correlation between entry width and CMRR (figure 6) [Mark

and Chase 1994]. The relationship between intersection span

and the incidence of roof falls at six mines was documented by

Mark et al. [1994] (figure 7). A similar correlation is reported

by Molinda et al. [2000].

FORCES APPLIED TO THE COAL MINE ROOF

Stress is everywhere underground (figure 8). Usually, the external forces applied to rock are all compressive, but they are not

equal in all directions. The in situ stresses are normally resolved

into three components: (1) vertical stress, (2) the maximum horizontal stress, and (3) the minimum horizontal stress.

Vertical Loads

The most obvious source of loads on mine structures is the

weight of the rock itself. It is convenient to analyze two types

of vertical loads (figure 9):

$ The roof load, which is due to the weight of the

immediate roof strata as they sag into the mine opening; and

$ The pillar loads, which are applied by the weight of the

overburden.

The roof load is the vertical force that most directly applies

to roof support. Various "dead weight" design methods are

based on estimating the volume of immediate roof rock that has

separated from the more stable overlying rock mass and consequently must be supported [Mark 2000].
28

Figure 6.—Entry widths and CMRR in U.S. longwall mines [Mark and Chase 1994].

Figure 7.—Relationship between CMRR, intersection span,

and roof fall rate at six U.S. mines [Mark et al. 1994].

Figure 8.—Stress on a typical element of mine roof.

The overburden load, on the other hand, is primarily carried

by the pillars, but it can affect the immediate roof stability (and

thus support loading) byC

$ Causing sloughage of the pillars, thereby increasing the

roof span in the mine entry;

$ Excessively loading or yielding the pillars or the mine

floor, resulting in differential movements that can damage the

immediate roof rock;

$ Stressing the pillars, which causes them to squeeze out

and apply a horizontal force to the immediate roof rock above

the mine entry.

Horizontal Stress

The horizontal stresses are normally more important to roof

stability than the vertical stresses. The reason is that most vertical stress is applied to the pillar, whereas the roof must bear

the full brunt of the horizontal stress. Moreover, the magnitude
29

Figure 9.—Vertical loads in underground coal mines.

Figure 10.—Stress fields in the continental United States [Zoback and Zoback 1989].

of the horizontal stress is usually greater than the vertical stress.

The effects of horizontal stress areC

$ Compressive-type roof failures (commonly called cutter

roof, guttering, shear, snap top, and pressure cutting). In thinly

bedded roof, the failure develops as the progressive layer-bylayer crushing of the individual beds.

$ Directional effects, because roof damage is generally

much greater in entries oriented perpendicular to the maximum

horizontal stress than in entries driven parallel with it.

During the past 15 years, horizontal stress has become

central to an understanding of coal mine ground control. An

important breakthrough was the recognition that the stresses

observed in coal mines are caused by global plate tectonic

forces [Mark 1991]. The World Stress Map Project [Zoback

and Zoback 1989] identified stress regimes in many parts of the

world by analyzing active faults, borehole breakouts, and

hydraulic fracturing stress measurements (figure 10).

An evaluation of stress measurements made in underground

coal mines confirmed that the stress map applies to underground

coal mines [Mark and Mucho 1994]. In the Eastern United

States, 76% of the measurements fell within 25E of N. 75E E. 
30

Figure 11.—Horizontal stresses in measured eastern U.S. coal mines [Mark and Mucho 1994].

Figure 12.—Horizontal stresses measured in western U.S. coal mines [Mark and Mucho 1994].

In magnitude, the horizontal stresses were generally two to

three times the vertical (figure 11). In the Western United

States, there seems to be much more variation from mine to

mine and even within individual mines. The horizontal stress

is also approximately equal in magnitude to the vertical stress

in the West (figure 12). In both sets of measurements, the

maximum horizontal stress was usually about 40% greater than

the minimum.
31

Figure 13.—Horizontal stress concentration in leading

entries [Mark and Mucho 1994].

Figure 14.—Horizontal stress concentrations in longwall

headgates [Mark and Mucho 1994].

Two other factors also determine the degree to which horizontal stress will affect ground control:

$ Roof type: Weak roof is more likely to suffer damage

than strong rock, and laminations or thin bedding (as in shales

or stackrock sandstones) greatly reduce the ability of rock to

resist horizontal stress.

$ Surface topography: Stream valleys can concentrate horizontal stresses and have often been associated with particularly

difficult horizontal stress conditions. Stream valleys can also

reorient the maximum horizontal stress away from the regional

direction [Molinda et al. 1991].

Stress measurements are too expensive for most mines to use

routinely. As a substitute, procedures have been developed to

estimate the orientation of the maximum principal stress [Mucho

and Mark 1994; Fabjanczyk 1996]. Such features as roof

"guttering" or roof "pots" are mapped underground, and the stress

direction is inferred from their orientation and severity.

Mining-Induced Stresses

The act of mining can concentrate and reorient the original

in situ stresses. Whenever coal is removed, the overburden

weight that it had carried is transferred. Vertical stresses are

therefore increased on the adjacent unmined coal. Horizontal

stresses are similarly affected when the roof is deformed or

fails. Horizontal stress cannot pass through broken ground, so

it becomes concentrated where the roof is still intact.

$ Development mining: Entry development creates pillar

loads, and "transient stress abutments" have been observed

[Karabin et al. 1982]. Horizontal stress creates more serious roof

control problems. In some mines, "leading entries" are heavily

damaged and require extensive support. Adjacent entries can be

stress relieved [Mark and Mucho 1994] (figure 13). Outby the

face, a roof fall can create a horizontal stress concentration, which

can then propagate itself hundreds of feet.
32

Figure 15.—Multiple-seam mining and its effects on ground

control [Mark 1990].

$ Retreat mining: Longwall mining and pillar recovery can

concentrate large vertical loads on gate entries and pillar lines.

Proper pillar sizing is essential for limiting the roof stresses and

deformations to levels that can be handled by roof support

[Mark and Chase 1994, 1997; Colwell et al. 1999]. Secondary

support is generally necessary to help control the additional

loads. Recently, the importance of horizontal stress abutments

has also been documented [Mark et al. 1998] (figure 14).

Proper panel layout can greatly reduce the loads applied to the

roof.

$ Multiple-seam mining: Overmining and undermining are

responsible for some of the most severe conditions found

underground. Both can concentrate vertical loads, and undermining can cause subsidence that damages the roof above overlying coalbeds (figure 15) [Chekan and Listak 1994].

THE FUNCTIONS OF ROOF SUPPORT

Support systems work best when they enhance the inherent

strength of the mine roof [Hoek and Wood 1988]. They can do

this byC

Providing confinement. Rock is much stronger when it

is confined. Since roof rock is usually being loaded by horizontal stress, even a small amount of vertical confinement can

have a big effect. The frictional strength of bedding planes may

also be strengthened by confinement.

Limiting deformation and preventing unraveling. By maintaining the integrity of the roof line, supports help the upper

layers maintain their strength.

Tying weaker rock units to stronger ones. Coal mine roof

often consists of several layers of rock with different strengths.

Roof bolts are particularly effective in tying weak or broken

rock to beds that are more self-supporting.

When the rock is completely broken and has lost all of its

strength, supports can also carry the dead-weight load.

PROPERTIES OF ROOF SUPPORT SYSTEMS

Roof supports can be divided into two categories:

$ Intrinsic support, where the supporting elements are

installed within the roof; and

$ Standing support, where the supporting members are

installed between the roof and floor.

Roof bolts are the best example of intrinsic supports. Roof

bolts are loaded as the roof deforms, and they interact with the

rock to reduce bed separation by confinement much as reinforcing steel does with concrete. Standing supports, like cribs,

posts, or longwall shields, develop loads in response to the

convergence between the roof and floor.

CAPACITY OF ROOF SUPPORTS

The first question usually asked regarding a support system

is: "How much load can the support carry; what is its capacity?" For roof bolts, two types of capacity are normally

given: the yield and the ultimate (figure 16). In general, these

can be calculated from the properties of the steel and the diameter of the bolt. However, as discussed by Mark [2000],

poor anchorage can substantially reduce the effective capacity

of roof bolts.

The capacities of standing supports depend on several factors, including the materials, configuration, and height. In

general, the capacity of each particular support type must be
33

K ' A ( E

L (1)

Figure 16.—Yield and ultimate strengths of a roof bolt.

Figure 17.—Example showing that the capacity of a

support should be defined in relation to its displacement.

determined by controlled load testing. The Pittsburgh Research

Laboratory of the National Institute for Occupational Safety and

Health (NIOSH) has tested a large number of supports in its

unique Mine Roof Simulator. From these tests, the

performance characteristics of these various support systems

have been determined. By matching the support characteristics

to the ground behavior, an optimum support design can be

achieved. To facilitate this approach, NIOSH developed the

Support Technology Optimization Program (STOP). This

program allows the user to determine the optimum installation

parameters for any support technology and compare the installation of one support system to another in terms of installed

support load density and the convergence control provided by

the support [Barczak 2000c].

In many cases, however, the capacity may not be the most

meaningful way to define a support. Consider the example shown

in figure 17. The second support (support No. 2) has twice the ultimate capacity of the first support, but it takes four times the convergence to reach this capacity. Furthermore, at one unit of displacement, the second support has only one-half the capacity of

the first support. Figure 18 is another example of the importance

of defining the support capacity in relation to the displacement.

Although this support has an ultimate capacity of >1,000 tons, is

that really meaningful? Before this capacity is mobilized, nearly

5 ft of convergence must occur. By that time, most entries would

be entirely unserviceable. Clearly, a better question is: "How

much load can the support carry at a specified amount of displacement?" This leads directly to the issue of support stiffness.

STIFFNESS OF ROOF SUPPORTS

Stiffness is simply a measure of how quickly a support develops its load-carrying capacity in response to convergence.

Stiffness is a measure of performance before a support reaches

its maximum capacity. Stiffer supports develop capacity more

quickly (with less displacement) than softer supports. The support elements can be thought of as large springs. A softer

spring will compress a greater amount to provide the same

resisting force as a stiffer spring. A good analogy is to think of

a 1⁄2-ton and 3/4-ton pickup truck. The 3/4-ton truck has stiffer

springs on the bed of the truck. Thus, if these two trucks were

placed side by side and each was loaded with a cord of firewood, the bed in the 1⁄2-ton truck would be lower than the bed

in the 3/4-ton truck (figure 19).

While some roof supports are installed with an initial preload, they all develop their load-carrying capacity only through

movement of the roof. This creates a fundamental paradox in

roof support design. The roof must deform to mobilize the support capacity, but it is this very movement that the support is

trying to prevent. Thus, a critical design issue is the stiffness of

the support system.

Since stiffness is such an important design parameter for

roof supports, let us examine some of the things that impact the

stiffness of a support structure. Stiffness (K) is a function of

the area (A), material modulus of elasticity (E), and the length

or height of the support (L), as expressed in equation 1.

Thus, as seen in equation 1, stiffness increases with area and

material modulus and decreases with increasing support height.

The significance of these parameters can best be understood by

looking at some practical examples. 

Intrinsic Support

Let us first examine the implication of these parameters on

roof bolt stiffness. First, since roof bolts are made from steel

and the modulus of elasticity of steel varies little, the stiffness
34

Figure 18.—A support that requires 5 ft of convergence to reach peak load.

Figure 19.—Pickup truck analogy illustrating support stiffness.

The heavy spring in the 3/4-ton truck deflects less than the light

spring in the 1⁄2-ton truck when both are loaded with the same cord

of firewood.

Figure 20.—The free length of a pointanchor roof bolt affects its stiffness.

of roof bolts is not affected by the grade of steel used in

fabricating the bolt. However, since the stiffness increases in

direct proportion to area or the square of the bolt diameter, bolt

stiffness increases dramatically with increasing bolt diameter.

Thus, a 7/8-in-diam bolt is twice as stiff as a 5/8-in-diam bolt,

all other things being equal.

Bolt length also affects stiffness. With a conventional pointanchor mechanical roof bolt (figure 20), the bolt is anchored

only at the top, and the "free length" of the bolt is defined as the

length of bolt below the anchor. Thus, as the bolt length increases, the stiffness of the bolt decreases, meaning that longer

bolts have a softer response and allow more roof movement to

occur for the same increase in bolt load. Fully grouted bolts, on

the other hand, do not initially have a "free length" and usually

become highly stressed in localized areas in response to roof

movements. For this reason, fully grouted bolts are normally

considered to be stiffer than point-anchor bolts. Cable bolts and

trusses are the least stiff of the intrinsic supports [Dolinar and

Martin 2000].

Standing Support

The same principles apply to standing support. Using wood

cribs as an example, 9-point cribs are stiffer than 4-point cribs

because the timber contact area of a 9-point crib is 2.25 times

that of a 4-point crib. Likewise, a 10-in-diam post will have a
35

Figure 21.—The stiffness of a wood crib support system

is increased by increasing the support density.

Figure 22.—The stiffness of a concrete crib is reduced by placing wood timbers on top.

stiffer response than a 6-in-diam post. Wood cribs can be made

stiffer by using different wood species. For example, the elastic

modulus of oak is greater than that of poplar wood; thus, oak

cribs will be stiffer supports than equivalent cribs constructed

from poplar timbers. The stiffness of standing supports is also

height-dependent, decreasing with increasing height. For example, a 4-point wood crib constructed from 6×6×30-in, mixed

hardwood timbers in a 6-ft seam height will provide 41 tons of

support capacity at 2 in of convergence, whereas the same crib

design constructed in a 10-ft seam will provide only 32 tons

(a 25% reduction) at 2 in of convergence.

Both intrinsic and standing support systems can made stiffer

by increasing the density of the supports. An example is shown

in figure 21, where two rows of wood cribs are increased to

three rows, with the middle row staggered with respect to the

two outer rows. Another approach to increase the system stiffness is to reduce the spacing between supports.

Supports can also be softened by adding additional material

on top of the support or within the support during its

construction. The rule to remember here is that of the weak-link

principle—the softest material will control the initial stiffness

of the support. The load-displacement response of a concrete

crib topped off with a row of wood timbers is shown in

figure 22. It is seen in this figure that the wood, which is the

softer of the two materials, controls the initial load development

of the support. The same principle applies to timber posts

where cap boards and/or wedges are used on top of the post.

Here, the material may be the same, but wood is much stronger

and stiffer when loaded parallel to the grain as in the post

section compared to perpendicular to the grain, as would be the

case for the cap blocks or wedging material.

RESIDUAL STRENGTH

What happens to a support after it reaches its maximum

capacity can be just as important as what happens before.

Consider the concrete crib constructed from concrete block

typically used in stopping walls and the 24-in-diam Can support

shown in figure 23A. Both have approximately the same initial

stiffness and capacity. However, once the concrete crib reaches

its maximum load, it fails completely, leaving the roof entirely
36

Figure 23.—Residual strength. A, The Can has better residual strength than the concrete block crib. B, The Propsetter has better

residual strength than the timber post. (Note: 1 kip ' 1,000 lb).
37

unsupported. The Can, on the other hand, continues to carry

nearly all of its load as the roof continues to move down as

much as 2 ft. A similar comparison can be made between a

conventional timber post and a Propsetter support (figure 23B).

The residual strength of supports like the Can and Propsetter

make them much more useful in moderate to high convergence

such as longwall tailgates than brittle supports like the

conventional concrete crib and timber post.

OTHER SUPPORT CHARACTERISTICS

Stability

Stability can be defined as the capability of a support to

sustain its load-carrying capacity through a useful range of

convergence without failing prematurely. Instability that results

in premature failure can be caused in several ways, the most

common of which are—

• Buckling, which is common in timber posts and most proptype supports (figure 24A);

• Material failure, where the load applied to the support

causes the material to fail in all or part of the support such that

the integrity of the support is compromised (figure 24B);

• Eccentric loading, which can be caused by wedging of the

support in place or uneven roof and floor contact (figure 24C);

and

• Lateral roof-to-floor loading, usually caused by differential

floor heave, which causes the support to lean or tilt off axis

(figure 24D) [Barczak 2000b].

Material Handling Requirements

Each year, 5,000 workdays are lost by workers in underground coal mines from timber handling injuries alone. In

recent years, new support technologies have been developed,

including engineered timber support systems, that dramatically

reduce the material handling requirements for standing roof

support systems [Barczak 2000a].

Installation Quality

In order to get the full benefit of the support, it must be installed properly. Improper installation of support is a major

cause of premature support failure. Each support is different,

thus the critical parameters for proper installation vary from

support to support. Some examples are—

• Wood cribs: The performance of wood cribs can be

degraded in several ways due to poor installation. For example,

the timbers should be overhung to allow the timbers to interlock

more effectively, thereby improving the crib stability during

loading (see figure 25A). Constructing the crib with the wide

side of the timber place up will reduce the capacity and degrade

the stability of the support. Rounded support timbers will also

reduce crib stability and capacity (figure 25B). If possible,

these timbers should be replaced by square timbers during the

construction process. Timbers should also be of consistent

quality. One weak or poor-quality timber can severely degrade

a 4-point wood crib, since each timber must function to provide

the full support capability (figure 25C).

• The Can: The Can Support is a thin-walled steel container

that is prefilled with air-entrained concrete before the unit is

transported into the mine. Proper installation requires a layer of

good-quality timbers that provides full coverage of the top of

the Can to preserve the design load profile. If this is not done,

the timbers will not have adequate strength to transfer the

loading to the Can; instead, the initial load profile of the support

will be unintentionally softened by the wood timber response

(figure 26).

• Roof Bolts: Obviously, roof bolts depend on proper

anchorage to achieve the rated bolt capacity. For grouted bolts,

proper mixing and hold time during the bolt installation are

critical. Grout performance is affected by several factors, including temperature, age, and conditions of storage.

Timing

Another way to define supports is by the time of installation.

Primary supports are installed immediately upon development.

In the United States, primary supports are almost always roof

bolts. Secondary supports are placed in anticipation of additional loading, as in a longwall tailgate. Supplemental supports

are used when the original supports are insufficient.

Skin Control

Skin control is the ability of a support system to prevent

injuries from small pieces of falling rock. With roof bolts, skin

control may be supplied by plates, headers, straps, or

mesh [Bauer and Dolinar 2000]. Skin control is also the reason

why many miners would prefer two rows of 4-point wood cribs

to a single row of 9-point cribs, even though the load-bearing

capacities are nearly the same for both support systems.
38

Figure 24.—Examples of support instability. A, buckling; B, material failure; C, eccentric loading; D, lateral

roof-to-floor movement.
39

Figure 25.—Examples of poor crib construction. A, rollout of crib blocks due to inadequate overhang; B, rounded timbers degrade

support; C, a single weak crib block causes premature failure of a "mixed hardwood" crib.

Figure 26.—Poor-quality timber on top of Can degrades support.

SUPPORT AND STRATA INTERACTION

The goal of roof support is to create a stable rock structure.

The properties of the roof and the magnitude of the rock

stresses determine the quantity of roof support that is required.

The support must also withstand the deformation that occurs in

the roof.

The concept of the "ground reaction curve" was developed

to illustrate the interaction between the load and the roof

movement [Scott 1989]. A ground reaction curve may be

defined as "the set of possible support loads required to achieve

stability for a given roof." The ground reaction curve depends

on the rock mass quality, the span, the in situ stress, and the

mining-induced stress. A change in any of these variables can

cause the ground reaction curve to shift, thereby increasing or

decreasing the support load required (figure 27).

The ground reaction curve forcefully shows that deformation, as well as load, is critical to proper roof support design.

The importance of support characteristics can be illustrated

using the ground reaction curve. If the support is too soft, it

may not be able to develop the necessary support capacity to

prevent excessive deformation from occurring (figure 28).

A support with little residual strength may fail prematurely if

the curve shifts because of additional mining stresses. Mucho

et al. [1999] describe how a tailgate ground reaction curve can

be measured and used to select the proper support density for a
40

Figure 27.—Ground reaction curves and the factors that affect them.

Figure 28.—Effect of support stiffness on the ground

reaction behavior. Figure 29.—Effect of installation timing.
41

particular support design. This capability is also provided by

the Support Technology Optimization Program (STOP).

As illustrated by the ground reaction curve, the "ideal" roof

support has the following properties:

• High initial stiffness, so that only small ground movements

are needed to mobilize the capacity of the support;

• Large load-bearing capacity; and

• High residual strength over a large range of displacement.

Many of the engineered timber and concrete supports have largely

succeeded in displaying these characteristics. Traditional wood

supports have somewhat less desirable characteristics. Simple

timber posts have little residual strength, while wood cribs have

a low initial stiffness.

Since passive supports must be compressed to develop their

load-carrying capacity, if they are installed too late, they might

not develop sufficient capacity in time to put the roof into

equilibrium. This is shown in figure 29. Both supports in this

example have the same stiffness, but the second support was not

installed in time to prevent critical roof deformation and thus

could not prevent a roof fall.

CONCLUSIONS

Roof supports work best when they are matched to the

ground conditions in which they are used. The performance

characteristic of each support is unique. A support system may

perform well in one application, but not in another. Understanding the ground, applied loads, and support characteristics

 are the keys to optimizing support design and application. The

goal of the papers in these Proceedings is to provide the best

available information and design guidelines to help mine

planners in this task.

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Australia: Coalfields Geology Council of New South Wales, pp. 137-142.

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43

TRENDS IN ROOF BOLT APPLICATION

By Dennis R. Dolinar1 and Suresh K. Bhatt1

ABSTRACT

The roof bolt system of a mine, if properly selected and installed, can allow for better roof control and

reduce the potential for roof falls. Because of this potential to reduce roof falls and improve ground control

conditions, the National Institute for Occupational Safety and Health (NIOSH) has an interest in the types of

roof bolts that are used by the coal industry. Further, NIOSH has conducted research into how various support

parameters can affect the number of roof falls that occur, including support type. Today, the five main types

of roof bolts installed in U.S. coal mines are mechanical anchor bolts, point-anchor or resin-assisted

mechanical anchor bolts, torque-tension bolts, combination bolts, and fully grouted resin rebar. This paper

describes each support in detail. Because of the importance of resin in the functioning of the majority of bolts

installed, resin grouts are also discussed. Further, trends in the types of roof bolts used are reviewed. Over

the last 10 years, the significant trend has been the large reduction in the relative number of mechanical anchor

bolts that are installed. These bolts have been replaced mainly by the resin-grouted rebar system. Data are

also presented to show the impact of these changes in bolting on the number of roof falls.

1

Mining engineer, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
44

Figure 1.—Percentage of bolt types used in U.S.

underground coal mines in 1999.

INTRODUCTION

Roof bolting in coal mines was initiated on a significant

level only after World War II. At that time, the technological

factor that drove the change from timber support to roof bolts

was the increased mechanization in the mining industry, while

the introduction of carbide-tipped drill bits made the drilling of

bolt holes on a production basis feasible [Thomas 1950, 1954].

The change to roof bolts accelerated during 1949 after the steel

shortage caused by the war had eased. Before 1949, only a few

coal operations had experimented with roof bolting. By May

1949, there were more than 200 mines, mostly coal operations,

that were installing roof bolts. Roof bolts supported over

14 million ft2

 of roof, using an estimated 800,000 to 1 million

roof bolts. By the end of 1949, more than 430 coal mines used

roof bolts.

Most of the bolts used a slot and wedge anchor, but some

mines were already experimenting with expansion-shell or mechanical anchor bolts [Thomas et al. 1949]. The slot and wedge

bolts had a threaded section at the head of the bolt and were

tensioned by a nut tightened against a bearing plate. Because

of the superior anchorage, the mechanical anchor bolts

eventually replaced the slot and wedge bolts as the main roof

support.

Even in the early days of roof bolting, it was recognized that

a support in full contact with the rock along its entire length

would be useful in dealing with rock shear forces, especially

along hanging walls of hard-rock mines [Thomas 1954]. For

the coal industry, the full contact support in the form of a fully

grouted rebar was introduced in the late 1960s to early 1970s.

This rebar bolt now is the predominant support used by the coal

industry. Resin-assisted mechanical anchor or specialty bolts

that combined the superior resin anchor with the tension of a

mechanical anchor bolt were introduced in the late 1980s.

Today, the five main types of roof bolts used in the U.S. coal

industry are fully grouted resin rebar, mechanical anchor bolts,

resin-assisted mechanical anchor bolts, torque-tension bolts,

and combination bolts.

The type of bolt can be important for roof control because

each support has characteristics that determine how the bolt will

support the roof. The main characteristics that differentiate

supports are whether the bolt is pretensioned and whether the

anchorage length is full or point contact. How the bolt will

interact with the site-specific rock mass properties and stress

conditions and stabilize the roof must be considered when

selecting a support [Mark 2000; Deere et al. 1970; Scott 1989].

The design of the support system must also consider other

support properties, including anchorage capacity, anchorage

load distribution, axial stiffness and toughness, shear resistance,

shear stiffness, and shear toughness [Karabin et al. 1980].

This paper discusses the recent trends in the types of bolts

used and the impact on roof control as the type of reinforcement

has changed. The main types of bolts used today are described,

and resin grouts used to anchor most of the rock bolt systems

are discussed.

TRENDS

To evaluate roof bolt usage and trends in U.S. coal mines,

the National Institute for Occupational Safety and Health

(NIOSH) collected information from all known U.S. roof bolt

manufacturers. Figure 1 shows the results of this study for

1999. However, the data were not complete in every instance,

and the data are for all bolt usage, not just coal mines. Therefore, the results shown in figure 1 should be considered

estimates. In the figure, five bolt types are indicated:

mechanical anchor, resin-assisted mechanical anchor, torquetension, combination, and fully grouted resin rebar bolts.

Although this figure does not break down bolt usage by

commodity, based on coal production, the coal industry uses an

estimated 80% to 85% of the reinforcement. Therefore, the

percentages of the different types of bolts used are probably

very representative of the distribution of the type of bolts used

in coal mines. In 1999, approximately 100 million bolts were

used in the U.S. mining industry.

Fully grouted resin rebar comprises about 80% of these

bolts. For the grouted rebar, approximately 80% are 0.625-indiam #5 rebar, and nearly all of the remaining are 0.75-in-diam

#6 rebar. Mechanical anchor bolts comprise about 8% of the

supports, while torque-tension bolts represent 6% of the

supports. Resin-assisted mechanical anchor bolts are approximately 5% and combination bolts about 1% of the market.

Surveys on bolt usage were also conducted in 1988 and 1991

[Scott 1989]. Table 1 shows a comparison between the percentages of each bolt type for these years and 1999. An

estimate of the distribution of bolt types for 1976 is also given

in the table. In that year, 80% of the bolts used were
45

Figure 2.—Incident rate for reportable roof falls based on

production from 1988 to 1998 for both longwall and room-and-pillar

operations, including a regression analysis.

mechanical anchor and 20% resin-grouted rebar [Karabin et al.

1980]. From 1976 to 1991 and from 1991 to 1999, there was

a substantial shift away from mechanical anchor bolts to fully

grouted resin rebar. For resin-assisted mechanical anchor bolts,

there was also a small decrease in the percentage used from

1991. In the 1991 survey, no distinction was made between

torque-tension and combination bolts; the two systems were

classified as point-anchor tension rebar. However, the combined usage of these two systems was about the same for 1988

and 1999, with a slight drop in the percentage used for 1991.

Table 1.—Bolt usage in U.S. mines by type

 Bolt type Percent

1976 1988 1991 1999

Mechanical anchor ......... 80 35.3 34.1 8

Fully grouted . . . . . . . . . . . . . . 20 40 48.2 80

Torque-tension . . . . . . . . . . . . — 3.5 1

 4.6 6

Resin-assisted mechanical

 anchor . . . . . . . . . . . . . . . . . . — 14.1 11.5 5

Combination . . . . . . . . . . . . . . — 3.5 — 1

Other1 — 3.6 1.5 —

 Total2 .................. 100.0 100.0 100.0 100.0 1

Includes both torque-tension and combination bolts.

2

Data may not add to totals shown because of independent rounding.

In general, resin-grouted rebar, despite being a passive support, may be considered a superior system to the mechanical

anchor bolt because of the anchorage capacity and load transfer

capabilities. Therefore, with the significant shift toward resin

bolts, a corresponding improvement in roof stability might be

expected that could be traced over the last 10 or even 25 years.

The number of reportable roof falls that occurs each year can be

used to evaluate whether a significant improvement has occurred.

Table 2 shows the number of reportable roof falls, the number of

mines, and the tons mined from 1989 to 1998 for longwall and

room-and-pillar mining. To better compare the data for each

year, the roof fall rate based on production is also shown. Figure

2 shows the roof fall rate per million tons of coal from 1989 to

1998 for both longwall and room-and-pillar mining.

For room-and-pillar mining, the roof fall rate trend from

1989 to 1998 can be evaluated by fitting a linear regression to

the data. The results indicate that the coefficient of determination is <0.1, while the slope of the line is not significantly

greater than zero. Essentially, there has been no change in the

roof fall rate for the last 10 years in room-and-pillar mines. For

1975 and 1976, the roof fall rate per million tons was 5.667 and

6.841, respectively. Compared to the rates for 1997 and 1998

(7.067 and 7.011, respectively), there certainly has not been any

decrease in the fall rate over 25 years.

For longwall mining, there was a decrease of about 1.6 roof

falls per million tons, or a reduction of about 50% in the roof

fall rate between 1988 and 1998. However, other factors such

as face width, seam height, and number of gate road entries

have caused much of this change by reducing the amount of

development mining. Over this period, there was a 26%

increase in panel width and a 7% increase in seam height

[Merritt 1991; Fiscor 1999]. Further, the number of entries for

a gate road has been reduced from 3.5 in 1990 to 3.1 in 1999,

a decrease of 11%. Therefore, the increased panel width and

seam height and the decrease in the gate road entries could

account for nearly all of the decease in the roof fall rate. In

1990, about 26% of the production from a longwall was from

development [Bhatt 1994]. In 1999, an estimated 15% to 18%

of production came from development. Essentially, the change

in roof support has had a minimal effect on the roof fall rate in

longwall mines.

Therefore, despite the significant change from mechanical

bolts to fully grouted rebar, there has been little change in the

roof fall rate. This does not mean that the roof bolt type does

not influence roof stability. There is documented evidence for

specific cases where mines have changed bolt type and increased roof stability [Karabin and Hoch 1980; Peacock 1986;

Stankus 1991]. However, there are many other aspects to the

design of a roof support system other than the support type, and

in many situations roof falls may have been prevented only with

the addition of supplemental support. Further, the fully grouted

bolts might have allowed mining under more difficult roof

conditions, and therefore an increase in the roof fall rate could

be expected that is balanced by the use of the fully grouted bolt.

Also, a general analysis of all U.S. coal mines indicates that a

number of factors may have changed, including the number of

mines, roof conditions, and the accuracy of reporting roof falls.

Further, there are reasons for changing the roof support other

than for ground control, including different requirements for

checking installation quality. Lastly, the mechanical anchor

bolt has been replaced by a fully grouted resin bolt system

using a #5 rebar in a 1-in-diam hole, and this may not be the

optimum resin rebar system from a support standpoint. A more

detailed discussion of the potential problems with the #5 rebar

system compared to a #6 rebar bolt is presented below.
46

Table 2.—Roof fall rate per million tons mined for room-and-pillar and longwall mines

Year

Room-and-pillar Longwall

Roof

falls Mines Tons,million

Rate,

falls per

million tons

Roof

falls Mines Tons,million

Rate,

falls per

million tons

1989 . . . . 1,945 1,669 253 7.7 397 70 134 3.0

1990 . . . . 1,875 1,659 266 7.0 470 76 154 3.1

1991 . . . . 1,898 1,482 249 7.6 472 76 155 3.1

1992 . . . . 1,726 1,338 242 7.1 452 77 161 2.8

1993 . . . . 1,418 1,197 212 6.7 335 75 136 2.5

1994 . . . . 1,496 1,136 221 6.8 348 71 175 2.0

1995 . . . . 1,333 979 209 6.4 501 70 187 2.7

1996 . . . . 1,653 874 214 7.7 460 66 197 2.3

1997 . . . . 1,569 883 222 7.1 307 57 198 1.6

1998 . . . . 1,528 828 218 7.0 285 55 201 1.4

Besides the bolt type, information was also obtained on the

lengths of bolts installed. For the resin-grouted rebar lengths,

30% were under 4 ft, 47% were 4 ft, 13% were 5 ft, 8% were

6 ft, and 2% were >6 ft. The average bolt length for fully

grouted rebar was 4.2 ft; for the mechanical anchor bolt, 4.1 ft;

for the resin-assisted mechanical anchor bolt, 5 ft; and for the

torque-tension bolt, 5.4 ft. It seems that the torque-tension and

resin-assisted mechanical anchor bolts systems may be used in

more difficult roof conditions where an increased length is also

required to control the roof. However, no data from other years

are available for comparison.

ROOF BOLTS

The five main types of roof bolts used in U.S. coal mines

can be classified by two criteria: (1) anchorage length and

(2) whether the support is installed with pretension [Scott 1989;

Peng 1998]. Both criteria affect how the support actually

functions in supporting the rock. From an anchorage standpoint, the bolts are either point- or full-contact anchors. For a

point-anchor system, the anchorage lengths are usually <2 ft

and include the mechanical anchor and resin-assisted mechanical anchor bolt. With the full-contact supports like the

fully grouted rebar system, most, if not all, of the support is in

contact with the rock. Besides anchorage, the full-contact provides reinforcement through resistance to rock movement. The

full-contact support includes the fully grouted and torquetension bolts. Although the combination bolts are only partially

grouted, the anchor section can be considered a full contact

support.

The other criterion is whether the system is installed with

tension and therefore applies an active force to the rock. The

tensioned support systems include the mechanical anchor and

point-anchor bolts, which rely to a large extent on the active

forces developed from bolt tensioning to provide reinforcement

to the roof. The fully grouted rebar bolt is a nontensioned system. Both the combination and torque-tension bolts are active

supports, yet have a full-contact anchor along a portion of their

length. Therefore, along the anchorage portion, these supports

resist movement similar to fully grouted rebar, with an active

component to clamp the roof similar to the point-anchor

systems.

MECHANICAL ANCHOR BOLTS

At one time, mechanical anchor bolts were the main roof

support used in the coal industry. One advantage of mechanical

anchor bolts is quick installation (usually <10 sec). Today,

however, the mechanical bolt has been to a large extent

displaced by other support systems, primarily fully grouted

rebar. Mechanical anchor bolts consist of a smooth headed bar

with a threaded anchor end. A mechanical shell anchor

attached to the threaded end of the bolt is used to anchor the

system. As the bolt is torqued, the force drives a plug against

the outer shell, which expands and is set with a radial force

against the rock (figure 3). Once the anchor is set, the bolt is

then tensioned. Bolt torque is required to set the anchor and

provide an active force to the rock for reinforcement.

The tension can be up to the yield of the steel or the

anchorage capacity of the system. However, the anchor must

be able to support high bolt loads with minimal displacement.

Therefore, the anchor is critical to the functioning and capacity

of this system. In general, there are two types of mechanical

shell anchors: a standard and a bail anchor [Karabin et al.

1976]. The standard anchor has fixed leafs and makes only

point contact with the rock. Because of this point contact, this

anchor is usually better in stronger rock. The bail or free leaf

anchor allows for almost total shell contact along the borehole

wall and is therefore usually better in softer rock. However,

because of variations in design of the shell anchors,
47

Figure 3.—Mechanical anchored bolt installed in roof. Figure 4.—Resin-assisted mechanism anchor bolt installed

in roof.

underground testing is necessary to establish the best system for

a particular rock.

Anchorage capacities up to 25,000 lb have been achieved,

although the rock strength will always control the anchorage

capacity. Therefore, the rock limits the amount of tension that

can be applied by the mechanical anchor bolt and, in part,

whether this load will be sustained or bleed off. Over time, the

tension may be reduced because of creep or failure of the rock

around the anchor and relaxation of the anchor threads.

Therefore, the mechanical anchor bolt system has usually been

installed in stronger roof rock or at least where the anchor is

placed in a good-quality rock.

RESIN-ASSISTED MECHANICAL ANCHOR BOLT

Resin-assisted mechanical anchor bolts are essentially mechanical anchor bolts that have been transformed with the

addition of a resin plug. Today, they are often called pointanchor bolts. These systems can be installed almost as fast as

the mechanical anchor systems. Because they have greater anchor stability and less tension bleedoff than can be achieved

with only the mechanical anchor, these bolts are used where

ground control conditions are less favorable.

Although there are several varieties of resin-assisted mechanical anchor bolts, in general, the system consists of a

headed bar (either deformed or smooth) with a threaded end for

attachment of the mechanical shell anchor (figure 4). The

mechanical anchor shells are designed with resin passages to

allow for the flow of the resin around and below the anchor.

The resin anchorage is usually established by a short cartridge

of fast-setting resin with an anchor length of 1 to 2 ft. Mixing

of the resin is achieved primarily by inserting the bolt and

anchor through the cartridge. Compression washers and shells

are also used with some systems to compress the resin and may

result in improved anchorage [Stankus 1991]. The mechanical

anchor allows for the immediate tensioning of the bolt, while

the resin stabilizes the anchorage capacity. Often, the bolt

tension is set at 70% or more of the yield of the bolts with

torque loads up to 250 to 300 ft-lbf depending on the strength

of the bolt. However, the amount of tension that can be applied

is still limited by the rock strength. With the resin anchor, the

tension bleedoff should be less than with a mechanical anchor

system, although some bleedoff can still occur because of resin

creep and anchor thread relaxation.

FULLY GROUTED REBAR BOLTS

The fully grouted bolt is now the main support used in U.S.

coal mines, with about 80% of the market. The system consists

of a headed rebar anchored with a full-length column of resin

obtained from a cartridge (figure 5). The system is usually considered nontensioned, although plate loads of several thousand

pounds may develop during installation [Karabin et al. 1976].

There are also special techniques that will allow for even higher

installed loads [Tadolini et al. 1991]. The system works because of superior anchorage and stiffness that develops as a

result of the full bolt length resin anchor. Pull tests show that

it usually takes less than 2-ft length of a resin anchor to achieve
48

Figure 5.—Fully grouted resin rebar installed in roof.

the capacity of the support, although the "anchorage factor"

depends on the rock strength and other installation parameters

[Mark 2000]. The high stiffness is accomplished due to the full

contact grout anchor and the ability of the resin grout annulus

to quickly transfer the loads developed in the system back into

the rock. Because of the superior bolt stiffness, significant

resistance to the rock movement will be developed both axially

and laterally. Loads developed along the bolt where roof

separations occur will be quickly transferred back into the rock

and the movement resisted and limited at these points.

Because the system works as a result of the load transfer that

develops between the support and the rock, the annulus

thickness of the resin grout, the distance between the bolt and

the rock, is important to the proper installation and functioning

of the system. The optimum system annulus as determined

from experimental investigations was found to be about

0.125 in [Karabin 1976; Gerdeen et al. 1977]. Therefore, the

optimum system for a 1-in hole is developed by using a 0.75-

in-diam #6 rebar. Using an annulus smaller than 0.125 in has

been limited by practical considerations, such as the thrust

capacity of the bolter, the variation in the rebar diameters, and

the potential resin loss [Campoli et al. 1999]. However,

systems are now available that allow for only a 0.0625-in

annulus that have overcome some of these problems, including

controlling the rebar diameter [Tadolini 1998]. Testing has

shown that effective load transfer and anchorage capacities can

be achieved with this smaller annulus. Further, the installation

forces, although higher with the smaller annulus, should be

within the capacity of most roof bolt machines. There are two

aspects to the performance of the system based on the resin

annulus size, the bolt installation, and load transfer along the

bolt. Both will be impacted negatively by a larger annulus. Of

the rebar systems, the #5 rebar installed in a 1-in-diam hole is

by far the most widely used bolt. However, the annulus for this

system is 0.1875 in, which is larger than the optimum annulus

thickness.

Even though fully grouted bolts are a passive system, the

bearing plates are a necessary part of the support system.

Without a bearing plate, a fully grouted bolt can still function

and resist rock movement. However, surface control is lost

where the bearing plate is the first support element in controlling surface or skin failure. About 98% of all ground fall

injuries are from the failure of the surface of the roof or rib, and

the plate helps protect the workers from injury due to failure of

the skin of the roof. Further, loads of over 20,000 lb have been

measured at the bearing plates, which indicates that the plates

can add significantly to the roof support [Tadolini et al. 1986].

The plate will help the bolt resist roof movement in the lower

2 ft of the roof and is therefore an important element in roof

reinforcement.

TORQUE-TENSION BOLT

A torque-tension bolt is essentially a resin-grouted rebar

system that is pretensioned on installation. This system

consists of a rebar with a threaded end at the head of the bolt.

A nut with a torque-delay mechanism is used to torque and

tension the bolt, with a resin column used to anchor the bolt

(figure 6). With a full-column anchor, two different speeds of

resin are used. In the upper portion of the anchor, there is a

fast-setting resin; in the lower portion, a slower setting resin.

The bolt is inserted, then rotated in the resin, when the upper

fast resin sets, a torque-delay mechanism on the nut breaks and

the nut is rotated up against the plate, tensioning the system.

The applied load is distributed over the lower portion of the

bolt containing the slow-set resin. When the slower resin sets,

the system will resist rock movement with the stiffness of a

fully grouted bolt and will further reinforce the lower roof with

an active clamping force. A variation of this system is to leave

the lower portion of the bolt ungrouted. If the grout column is

sufficiently long, this upper part of the system will reinforce the

roof similar to a full-grouted rebar, with the lower portion of

the roof reinforced by the clamping action of the applied force.

However, the lower portion of the bolt will have much less

stiffness than the full-grout column and therefore less resistence

to rock movement. In general, the fully grouted torque-tension

system combines both the active force of the resin-assisted

mechanical anchor bolts and the superior anchorage and

stiffness of the resin rebar systems.

COMBINATION BOLTS

A combination bolt consists of a rebar anchor usually 3 to

4 ft long connected with a coupler to a smooth headed bar

(figure 7). The rebar section is anchored in the hole with a
49

Figure 6.—Torque-tensioned bolt installed in roof.

Figure 7.—Combination bolt installed in roof.

Figure 8.—Resin cartridge used to develop anchorage for

roof bolts.

resin cartridge. A shear pin in the coupler allows the resin to be

mixed and set before the lower smooth headed bar is torqued

and the system tensioned. Essentially, the rebar is a fullcolumn resin bolt that provides an anchor for the system and

provides reinforcement to the roof. With the tensioning of the

lower section of the system, a clamping force is applied to the

rock. These systems are often up to 8 ft long, and because the

system consists of at least two components joined by coupler,

a support system much longer that the seam height can be

installed with relative ease. However, the weakness of the

system is the coupler. Although the coupler is designed to

withstand axial forces up to bolt failure, coupler failure can

occur when there is sufficient lateral roof movement that causes

coupler failure by shear.

RESIN

Grout anchors used with coal mine supports are commonly

made with a polyester resin and packaged in a cartridge form.

These systems can be either water- or oil-based. The trend

today is toward water-based resins due to market forces. In the

cartridge there are three components, but only two are active.

These three components are the resin and the catalyst, which

are active, and an inert filler (figure 8). Usually, the filler

comprises between 65% to 75%, the resin 20% to 30%, and the

catalyst 2% to 3% of the system [Eaton 1993]. However, in the

United States some of the resin systems can have up to 85%

filler. The resin consists of a polyester polymer solid with a

liquid styrene monomer, while the catalyst is benzoyl peroxide.

The filler is usually a limestone and acts not only to fill the

hole, but also to help form the mechanical interlock between the

rock, bolt, and grout. The cartridge is formed with mylar packaging and is set up with two compartments to keep the resin

and catalyst apart. The packaging is torn and the system mixed

during insertion and rotation of the bolt during installation.

Two important functional parameters of the resins are the

mix time and the set or gel time. The mix time is usually

between 3 to 10 sec and is limited by the gel point. However,
50

proper mixing of the resin is based on a certain number of

rotations of the bolt that will occur during installation prior to

the gel point of the resin. The gel time is controlled by inhibitors added to the resin and is the time required for the resin

catalyst to change from a liquid to a solid; it will usually vary

from 5 sec to 1 hr or more. The gel time must be long enough

to allow for the installation of the support, but it is affected by

the rock, bolt, and resin cartridge temperature. The resin systems come with manufacturer-recommended installation procedures, which, if not followed properly, could result in a substandard anchor. Also, the annulus thickness will affect how

well the resin is mixed and therefore the anchor performance.

Strength and stiffness of the resin may affect the performance

of the anchor system. A higher elastic modulus will allow for

more efficient load transfer, while the higher strength will allow

load transfer to take place over a more extended range of bolt

load. The U.K. mining industry uses resins with strengths of

about 11,600 psi and an elastic modulus of 1.6 million psi. These

resins are designed to maximize anchor performance. In the

United Kingdom, it is believed that a resin grout should be as

strong as the strata that are being supported. In the United States,

lower strength resins are used, normally with a strength of about

5,000 psi. With these weaker and less stiff grouts, load transfer

may not be as effective and the grouts can fail sooner, resulting

in less overall load transfer. However, it is uncertain whether

stronger, stiffer resins would have a significant impact on roof

conditions in most mining situations. In many cases, the resins

used in U.S. coal mines are developing adequate anchorage for

load transfer and are able to withstand strata movement and

failure. The American Society for Testing and Materials

(ASTM) has also developed standards for resin testing, including

strength and mix time [ASTM 1998].

#5 VERSUS #6 REBAR IN A 1-IN-DIAMETER HOLE

The optimum annulus thickness based on experimental

evidence and practical considerations has been found to be

0.125 in [Karabin et al. 1976; Geredeen 1977; Campoli 1999].

In a 1-in hole, this annulus is achieved with a #6 rebar. However the annulus thickness for a #5 rebar is 0.1875 in, and there

is 30% less steel in the hole with this system. How and to what

degree the increase in annulus and decrease in steel affects the

performance of the system merits attention, especially since the

#5 rebar represents about 65% of support installed.

One way to compensate for the decrease in bolt diameter is to

increase the strength of the steel used for the #5 rebar. For many

mines, the grade 60 #5 rebar replaced a grade 40 #6 rebar.

Therefore, for the grade 60 #5 rebar the minimum yield load is

18,600 lbf compared to a grade 40 #6 rebar of 17,600 lbf.

Essentially, both systems have about the same yield load. This

increase in strength is accompanied by a loss of ductility in the

steel. The elongation of the grade 60 rebar is 9% at failure and

for the grade 40 rebar 12%. Further, the ultimate load for the #5

grade 60 rebar is 27,900 lbf, while the #6 grade 40 rebar is

30,800 lbf. The decrease in elongation and ultimate load results

in a significant loss of toughness or the ability of the #5 rebar to

absorb energy. Higher steel strengths such as grade 75, 90, and

100 are now used, while alloys can increase the elongation range

of the steel. However, the general problem of replacing the bolt

diameter with higher strength steels is still the loss of elongation

and therefore toughness of the system.

Both axial and lateral stiffness of the system is affected by

the annulus thickness. In tests conducted on 2-ft-long fully

grouted bolts in a 1-in-hole, the axial stiffness for the #5 rebar

was 142,000 lbf/in and for the #6 rebar 275,000 lbf/in [Bartels

et al. 1985]. In testing these systems in shear (across the bolt

axis), the shear stiffness for the #5 rebar was 30,200 lbf/in and

for the #6 rebar 131,000 lbf/in. The results of these tests show

a significant decrease in the stiffness both axially and laterally

of the #5 rebar system. The fully grouted bolt relies on the

system stiffness to resist rock movement both axially and

laterally. Therefore, a decrease in stiffness should impact the

support's ability to provide roof reinforcement.

With the larger annulus, more installation problems could be

expected mainly in the form of increased glove fingering. To

prevent this from occurring, extra care may be required to

ensure that the manufacturer's recommendations on spin and

rotation are followed during installation. Glove fingering will

affect the bond between the grout and the rock developed by

mechanical interlock and therefore the load transfer between

the support and the rock. Further, glove fingering will often

occur at the end of the bolt; thus, the anchorage capacity will be

reduced along a critical portion of the support system.

Essentially, the effective length of the support is reduced

because of inadequate anchorage at the end of the bolt.

Unfortunately, a pull test on a full-column grouted bolt will not

normally reveal this condition.

In many situations, the #5 rebar system seems to be an

effective support. However, some situations may require the

increased stiffness and load transfer of the more optimum #6

rebar system. Also, the effects of inadequate installation

resulting from the larger annulus have not been documented.

To date, as far as the authors know, there is no published

information on a direct comparison of the two systems in an

underground setting or even on any extensive laboratory studies

of the #5 rebar system. Therefore, the performance of the #5

rebar system needs to be evaluated in detail to determine if

there is any reduction in reinforcement capabilities over those

of the more optimum #6 rebar in a 1-in-diam hole. This is

especially important since the #5 system is by far the most

widely used support system in U.S. coal mines.
51

SUMMARY AND CONCLUSIONS

Today, there are five main types of roof bolts used in U.S.

coal mines: fully grouted resin rebar, mechanical anchor bolts,

resin-assisted mechanical anchor bolts, torque-tension bolts,

and combination bolts. The trends in bolt usage indicate that

the fully grouted rebar is the most widely used support,

representing nearly 80% of the market. Essentially, the fully

grouted rebar has replaced the mechanical anchor bolt where

mechanical anchor bolts are about 8% of the market. The fully

grouted bolt system is generally regarded as a superior reinforcement system compared to the mechanical anchor bolt,

and there are a number of documented cases where a change

from the mechanical anchor to the fully grouted bolt has

increased roof stability at individual mines. However, the trend

away from the mechanical anchor bolt to the fully grouted bolt

has not resulted in any noticeable change in the rate of

reportable roof falls in U.S. coal mines. It is possible that the

fully grouted bolt has allowed for mining under more difficult

roof conditions, or roof conditions in general have become less

favorable in the overall mine population. Under such conditions, a higher roof fall rate would be expected where the use

of the fully grouted bolt has actually offset this increase.

However, such a change in the overall roof conditions, if it has

occurred, might be difficult to document and prove. Further,

there may be other reasons besides improved roof stability to

change bolting systems, such as the different requirements to

check the quality of the installation of each system. The

torque-tension checks are more demanding for the mechanical

anchor bolt than the installation checks for a fully grouted bolt.

For the fully grouted bolt, the #5 rebar is the most widely

used and comprises nearly 65% of all support installed.

However, this is not an optimum system when compared to a

#6 rebar in a 1-in hole. Further, the strength and stiffness of

resins used by the U.S. industry are generally less than those

used in the United Kingdom and Australia. Therefore, how the

#5 rebar and the property of the resins have impacted roof

support performance is certainly open to discussion and further

evaluation based on the reportable roof fall data. However, the

selection of a specific bolt system to reinforce the roof is only

one aspect of the design of primary roof support systems,

although an important aspect for roof control.

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Thomas E [1950]. Roof bolting in the United States. Washington, DC:

U.S. Department of the Interior, Bureau of Mines, IC 7583.

Thomas E [1954]. The application of rock bolting as a means of ground

support is limited only by the resourcefulness and ingenuity of the mining

engineer. Min Eng, November.

Thomas E, Barry AJ, Metcalfe A [1949]. Suspension roof support.

Washington, DC: U.S. Department of the Interior, Bureau of Mines, IC 7533.
53

ASSESSING COAL MINE ROOF STABILITY THROUGH

ROOF FALL ANALYSIS

By Gregory M. Molinda,1 Christopher Mark, Ph.D.,2 and Dennis Dolinar3

ABSTRACT

In 1999, 2,087 unplanned roof falls were reported from 841 mines. Nearly 55% of all mines reported at

least one roof fall, and nearly 17% of the mines reported five or more falls. In order to investigate the variables

that contribute to roof falls, the National Institute for Occupational Safety and Health (NIOSH) compiled a

national database of roof performance from 37 coal mines. Geotechnical factors and their effect on roof fall

rates were compiled from over 1,500 miles of drivage. The factor that is the best predictor of roof fall rate is

the Coal Mine Roof Rating (CMRR). For a low CMRR (#30), almost all cases have high roof fall rates.

Conversely, high roof fall rates are rare for strong roof rocks (CMRR $ 60). Roof fall rates were also higher

in deeper mines, probably because of greater stresses. Intersections were much more likely to fall than

roadways, and four-way intersections were more prone to fall than three-way intersections. In a controlled

comparison of the effect of increasing bolt length on roof fall rates, it was found that longer bolts reduced the

roof fall rates in 11 of 13 cases. A relationship between the roof fall rate, the intersection span, and the CMRR

was also found. Finally, a systematic method for tracking roof performance and geotechnical variables was

demonstrated.

1

Research geologist.

2

Supervisory physical scientist.

3

Mining engineer.

Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
54

INTRODUCTION

In 1998, a total of 2,232 unplanned reportable roof falls

occurred in 884 U.S. underground coal mines. These falls

resulted in 419 injuries and 13 fatalities. According to the Mine

Safety and Health Administration’s (MSHA) accident database,

in 1999 over 55% of underground coal mines reported at least

one roof fall and 17% of the mines reported more than five

falls. In 1998, an estimated 12,500 miles of entry and 350,000

intersections were excavated. Falls of roof represent a very

small proportion of exposed and supported ground.

Nevertheless, each roof fall represents a direct threat to life and

limb or an indirect threat to ventilation, escape, and equipment.

Through trial and error, operators have generally learned

how to mine the coal and support the roof. After mining a

certain length of time in a given coal seam, the appropriate

entry width, mining height, length of cut, pillar geometry, and

support can usually be determined. Roof instability occurs

when conditions (usually geology, equipment, or economics)

change, and the operator is uncertain how to respond or adapt.

Past studies that focused on detailed measurements at

specific field sites have provided a wealth of data on specific

roof stability topics, including bolt loads, bed separation, rock

strength, mining influences, horizontal stress, and pillar stability

[Signer 1998; Dolinar 1997; Chase 1999; Wang 1996]. This

approach has been successful in increasing our understanding

of the mechanics of roof instability and failure. However, sitespecific instrumentation studies have the disadvantage that

measurements from that site may not be entirely representative

because of local variations in stress, geology, or support

installation.

Roof falls, after all, are relatively infrequent events. It is

difficult for deterministic rock mechanics models to explain

why one intersection collapses while many others nearby

remain stable. On the other hand, roof falls seem well suited

for study using a probabilistic or empirical approach. The basic

concept of the empirical approach is to collect a large quantity

of real-world case histories and then use statistics to determine

the most important factors.

The empirical approach also requires that the researcher

begin with a clear hypothesis, often in the form of a simplified

model of the real world that abstracts and isolates the factors

that are deemed to be important. It therefore requires, as

Salamon [1989] pointed out, "a reasonably clear understanding

of the physical phenomenon in question." Without prudent

simplification, the complexity of the problem will overwhelm

the method’s ability to discern relationships among the variables. But a key advantage is that critical variables may be

included even if they are difficult to measure directly through

the use of "rating scales."

During the past 5 years, modern empirical techniques have

been applied to a variety of problems in coal mine ground

control. They have resulted in some very successful design

techniques, particularly in the area of pillar design, as well as

some new insights into pillar and rock mass behavior [Mark

1999a, 1999b].

Much can be learned by observation of roof instability. The

geometry, timing, geology, and frequency of roof falls may

indicate what caused the failure. If these variables are carefully

documented on a mine-wide basis, it may be possible to

characterize the combination of factors that may contribute to

a high incidence of roof falls. This documentation, expanded

with corehole data, at a single mine is called hazard mapping.

The hazard map indicates that poor ground conditions are

expected. Responses can include bolt changes, narrowing the

span, and supplemental support in critical intersections.

Due to highly variable geology and stress regimes, it has

been difficult to transfer this knowledge to other mines. To

address this problem, the National Institute for Occupational

Safety and Health (NIOSH) decided to capture the experience

of thousands of miles of existing mine roadways to assess the

parameters that influence roof stability.

NATIONAL ROOF FALL DATABASE

Several geotechnical variables are known to influence roof

stability. These include geology, mine opening geometry,

horizontal and vertical stress regime, abutment load, and support. Through extensive interviews and underground reconnaissance with mine operators, NIOSH documented many of these

variables. A national database of roof falls was created from

data obtained during visits to U.S. coal mines (table 1).

Ultimately, 41 mines in 10 States were visited, representing

over 1,500 miles of drivage in most of the major coal basins

where underground mining occurs (figure 1). Study mines

were selected by computing the roof fall rate from the MSHA

accident database. Drivage was estimated by converting annual

production (excluding longwall production) into linear feet of

advance, assuming an average seam height. Reportable roof

falls were then divided by drivage to arrive at the roof fall rate

(figure 2). Mines were then selected for study from this

distribution to represent the entire range of roof stability from

high, to medium, to low roof fall rates. Mines were also

selected to represent a wide range of roof geologies, as well as

varying size, ranging from large (>1 million tons per year) to

small mines (<200,000 tons per year).
55
56
57
58

Figure 1.—Location of study mines.

 Figure 2.—Distribution of roof fall rates in U.S. coal mines.

At each mine, one or more "case histories" were collected.

A case history was a portion of the mine that could be defined by

a number of descriptive parameters and an outcome parameter (roof

fall rate, falls/10,000 ft of drivage). The outcome parameter was

based on the number of reportable roof falls that occurred in that

portion of the mine. 

According to MSHA regulations at 30 CFR 75.223, a fall of

roof is reportable when itC

$ Causes injury;

$ Falls above anchorage;

$ Blocks ventilation;

$ Stops production for 30 min; or

$ Blocks escape.

It was recognized that not all roof falls should be treated

equally because their causes and impacts vary widely. A protocol for filtering roof falls for the study was developed.

Tabulated roof falls were restricted to falls less than

18 months old in order to reduce time-dependent effects. Additionally, some mined areas are only accessible for short

times (retreat panels or gate roads). To ensure equal treatment, mined areas had to be open a minimum of 18 months

for use in the study.

Falls that were associated with longwall recovery, pillaring,

multiple-seam effects, or other abutment pressures were also

excluded. Falls associated with large-scale geologic dis-continuities, such as faults or sandstone channel margins, were

excluded because they represent anomalous conditions and

require specialized primary or supplemental support. In any

study of roof safety or support performance, falls due to these

factors must be treated separately because roof stability will not

be achieved by standard support practices.

Because geotechnical parameters vary within mines, it was

often not possible to characterize a whole mine by one set of

variables. As a result, it was possible to have two or more

"cases" within a mine representing a combination of geotechnical variables. The database ultimately included information from 37 of the 41 mines, but actually contained 109

"cases." The changing geotechnical environment of a mine roof

was characterized by partitioning sections of a mine into zones

with common variables. For example, a single mine might be

broken up into three zones, or cases, if three roof bolt lengths

were used. If two different roof geologies with different

CMRR values were encountered within each of the three bolt

length zones, then six cases were created.

ROOF FALL RATE

The roof fall rate was calculated as the outcome variable for

each case. It was calculated by dividing the total number of

roof falls that qualified for the study by the drivage. In order to

quantify the percentage of drivage affected by a roof fall, roof

falls were counted not as single entities but by the number of

intersections and entry segments involved. A single roof fall

covering two intersections and the crosscut between would

count as three falls. Figure 3 shows the distribution of roof fall

rates for the database. Nearly 60% of the cases in the data set

had no roof falls. The other outcome variable was the four-way

intersection rate. This number is calculated by dividing the

number of four-way falls by the total number of four-way

intersections in each case.
59

PRSUP ' Lb ( Nb ( C

Sb ( We , (1)

 Figure 3.—Distribution of roof fall rates for cases in the study sample.

PRIMARY ROOF SUPPORT

For this study, a careful effort was made to characterize the

roof bolts used in each case history. Operators were asked to

report what bolts were installed historically through the mine.

Where underground access permitted, NIOSH checked the

accuracy of the information by reading the roof bolt heads,

using a wire brush to clean them as necessary. This was done

routinely where access permitted. After underground verification, roof bolt maps were compiled for the entire mine (figure 4). Six bolt variables were documented for each type of

bolt used:

$ Bolt length.

$ Tension.

$ Length of grout column.

$ Yield capacity (grade of steel times cross-sectional area

 of the roof bolt).

$ Bolts per row.

$ Row spacing.

The most common bolt length used at the mines in our study

was 5 ft (figure 5). Over 3.2 million feet of drivage was supported by 5-ft bolts (38%). Six-foot bolts were the next

common length used (2.4 million feet of drivage, 30%),

followed by 4-ft bolts (1.83 million feet of drivage, 22%).

Bolt tension was defined as either tensioned or untensioned.

Eighty percent of the drivage was untensioned bolts (6.5 million

feet), and 20% of the bolts were tensioned (1.6 million feet)

(figure 6). All untensioned bolts were fully grouted, while

nearly all tensioned bolts were ungrouted or partially grouted.

Figure 7 shows the distribution of roof bolts in the study by

grout column and roof fall rate. The percentage of fully grouted

bolts far outweighs the other grout column types, mirroring the

national trend [Dolinar and Bhatt 2000]. There seems to be no

correlation between roof fall rate and grout column length (roof

fall rates are evenly distributed between variables), indicating

that other factors are involved in the roof fall rate. Figure 8

shows the distribution of tensioned bolts as related by roof fall

rate. Again, there is no correlation between roof fall rate and

tension. Yield capacity ranged from 8.8 to 22.5 tons, with

9.5 tons of capacity occurring most frequently. The pattern of

bolting in the United States varies little, with four bolts per row

across the entry and 4 to 5 ft spacing between rows standard in

nearly every case.

A summary variable, PRSUP, was calculated as a rough

measure of roof bolt density:

where Lb ' length of the bolt, ft;

Nb ' number of bolts per row;

C ' capacity, kips;

Sb ' spacing between rows of bolts, ft; and

We ' entry width, ft.
60

Figure 4.—Roof bolt map for study mine.

 Figure 5.—Bolt length distribution in database as normalized

by drivage.

 Figure 6.—Distribution of tensioned and untensioned bolts in

the study.
61

 Figure 7.—Relationship between bolt grout column length and

roof fall rate. Figure 8.—Relationship between bolt tension and roof fall rate.

Figure 9.—Relationship between PRSUP and roof fall rate.

Figure 9 shows the distribution of PRSUP of all cases as

grouped by roof fall rate. PRSUP is a rough measure of the

"intensity" of the support. The more "steel" in the roof, the

higher the PRSUP. It does not consider the type of bolt.

PRSUP differs from the PSUP used in past studies [Mark et al.

1994] in that the bolt capacity has been substituted for the bolt

diameter. The proportion of cases with high roof fall rates

increases with increasing PRSUP. Additionally, the average

PRSUP for cases with a roof fall rate equal to 0 is 4.4. The

average PRSUP for cases with roof fall rate >1 is 6.1 This

difference is significant at the % ' 0.05 level. This is an

indication that operators are responding to poor roof conditions

(higher roof fall rates) by adding more roof bolt support

(increasing PRSUP) and that they are being only partially

successful. The correlation between higher support densities

and higher fall rates also presented a problem for the statistical

study.

ROOF GEOLOGY

Roof geology has historically been difficult to quantify in

ground control studies because of the many factors that comprise it. The Coal Mine Roof Rating (CMRR) was designed to

quantify the geotechnical elements of the roof and return a

number from 0 to 100 that reflects the competence of the roof

[Molinda 1994]. The CMRR for each case was determined

primarily by underground observation of roof falls, supplemented by drill core when it was available. A roof geologyCMRR map was constructed for each mine (figure 10). Figure

11 shows the distribution of the CMRR in the database. There

is a strong correlation between CMRR and roof fall rate, with

higher roof fall rates in the weaker roofs (CMRR # 50). For

cases with a CMRR # 30, all have high or moderate roof fall

rates. Conversely, high roof fall rates are rare for roof rocks

with CMRR $ 60. If just the cases with no roof falls at all are

considered (n ' 41), the average CMRR is 52.3. For cases with

a roof fall rate $ 2.0 (n ' 36), the average CMRR is significantly lower at 42.8.
62

Figure 10.—CMRR map of study mine.

Figure 11.—Relationship between roof fall rate and CMRR.
63

Figure 12.—Method of measuring intersection diagonals.

 Figure 13.—Relationship between roof fall rate and intersection

span.

 Figure 14.—Comparison of the roof fall rate between intersections and entry segments.

 Figure 15.—Comparison between the roof fall rate for four-way

and three-way intersections.

INTERSECTION SPAN

The intersection diagonals were measured for a sampling of

intersections in each mine, and in cleaned-up roof falls when

possible. Figure 12 shows the method of measurement of

intersections. Measurements were averaged to represent the

typical span for each case. Figure 13 shows the distribution of

intersection spans and roof fall rate. There is no obvious

correlation between roof fall rate and intersection span.

While most (62%) of the falls in the total database occur in

intersections, the intersection fall rate shows that intersections

are much more likely to fall than entry or crosscut segments

between intersections (figure 14). Segments are defined as any

mined room that is not an intersection. Segments usually are

two to three times as long as intersections. Of the intersections,

four-way intersections are more likely to fall than three-way

intersections (figure 15).
64

Figure 16.—Relationship between roof fall rate and overburden.

DEPTH OF COVER

The depth of cover over the mines in the study ranged from

0 to 1,600 ft. For analysis purposes, the average cover was

recorded into three categories: shallow (0 to 400 ft), moderate

(400 to 800 ft), and deep (>800 ft). Figure 16 shows the

distribution of depth of cover for study cases.

STATISTICAL ANALYSIS

In order to determine the influence of the collected

geotechnical variables on roof instability, the database was

standardized and prepared for analysis. SPSS was the statistical

package used for the analysis.

One goal of the study was to determine if there was a

universal design equation that would use all or some of the

geotechnical variables to predict the roof fall rate. A linear

regression was performed that included all the significant

geotechnical variables, including overburden, bolt length, grout

length, density, entry width, CMRR, intersection span, tension,

and bolt capacity. The regression technique progressively

removes variables that are not significant in a stepwise

procedure. The resultant regression equation can explain only

29.9% of the variation of the four-way intersection fall rate.

More importantly, there is a positive relationship between bolt

capacity and the roof fall rate. In other words, when bolt

capacity goes up, the roof fall rate goes up. This defies logic,

but the explanation is that when roof conditions deteriorate

(roof fall rate goes up), higher capacity bolts are generally

installed.

There are several explanations for the low overall correlation

of the regression equation to the data. A test for intercorrelation

of the variables revealed that a number of the variables were

correlated to each other. This interdependence reduces the

overall correlation of the design equation. Table 2 (Pearson

correlation) is a test of the codependence of the geotechnical

variables. A value of 1.0 is a perfect correlation, and 0.0 shows

no correlation at all. Several bolt parametersCtension and grout

indices, capacity and density, and bolt tension and capacityCare

related. As roof conditions worsen, operators generally move

toward tensioned bolts as well as increased capacity, apparently

with only partial success. Intersection span and entry width are

naturally related. The CMRR and the bolt length are also

related. As expected, as the roof gets stronger (higher CMRR),

operators are installing shorter bolts. These intercorrelations of

variables confound the overall effect of any one variable on the

outcome, which is the roof fall rate. Therein lies the difficulty

in producing a reliable roof bolt design equation.

Although a universal design equation was not possible, the

data analysis produced other interesting results. Other studies

show significant evidence of increasing horizontal stress with

depth [Mark 1994]. In this study, there is indirect evidence of

the relationship. The data show that there is a statistically

significant correlation (Pearson correlation ' 0.253, statistically

significant at 0.01 level) between CMRR and overburden. It

seems that stronger roof rocks are encountered as overburden

increases. There is no geologic reason for this, but it seems that

operators are unable to mine weak roof at great depth. As

overburden increases, stronger roof is encountered. In our

database, 10 cases are mining at depths below 800 ft of cover,

and 9 are >50 CMRR.

Figure 17 shows the relationship between CMRR and depth

of cover for the study data. The individual cases have been

divided into three roof fall rate categories; high, borderline, and

zero. A line has been drawn on the graph that roughly separates

lower roof fall rates from higher roof fall rates. Sixteen of

twenty-two cases of zero roof falls fell above the classification

line and were correctly classified. Nineteen of twenty-three

cases with high roof fall rate (>2.0 falls per 10,000 ft of

drivage) fell below the classification line and were correctly

classified. The overall correct classification rate was 77%.

It seems likely that in our data, depth of cover is an indirect

measure, or surrogate, for horizontal stress level. Horizontal

stresses are seldom measured directly because of the difficulty

and expense.

Using this assumption, the case histories were divided into

two groups by depth of cover. The shallow-cover group

included depths <400 ft, and the deeper-cover groups included

depths $400 ft. Figure 18 shows the relationship between

CMRR and PRSUP at high cover. Ten of sixteen "high" roof
65
66

Figure 17.—Relationship between CMRR, depth of cover, and roof fall rate.

Figure 18.—Relationship between PRSUP, CMRR, and roof fall rate for cases under cover $400

ft.
67

Figure 19.—Relationship between PRSUP, CMRR, and roof fall rate for cases under <400 ft.

fall rates fell below the classification line and were correctly

classified. Eleven of 17 zero roof fall rate cases fell above the

classification and were correctly classified. Looking at the

"low" cover group, 16 of 17 cases with zero roof falls fell above

the classification line, whereas the high roof fall rate cases were

approximately evenly split (figure 19).

In both groups, most of the misclassified high-rate cases and

the borderline cases plotted fairly close to the classification

line. There were also high-roof-fall-rate cases that were misclassified that were also low CMRR. However, it seems that

the relationship may break down for the weakest roof.

The relationship between CMRR and PRSUP was used to

develop design equations, which are described by Mark

[2000].

OTHER VARIABLES AFFECTING ROOF FALL RATE

The data collected during this study showed a considerable

amount of scatter, as evidenced in many of the figures presented

thus far. The explanation for the scatter is that the mining

environment is far from a controlled experiment where all

variables may be held constant and varied individually. If this

were the case, the change in outcome variable can be observed

and attributed to one variable. Moreover, some variables are

difficult to measure, particularly over large areas. As a result,

the observed roof fall rates may be affected by a number of

factors that could not be included in the analysis, includingC

Geologic variation: Typically, in an underground coal mine,

parameters like geology (CMRR) can vary rapidly. Without

systematic roof exposure (test holes), it may be difficult to

assign the CMRR accurately to large sections of mine roof. By

underground observation of roof falls and other exposures and

drill core data, a CMRR was calculated that best represents the

case area.

Horizontal stress: The presence of high biaxial horizontal

stress is known to affect roof quality adversely [Mucho 1995].

The study used overburden depth as a surrogate for horizontal
68

stress, but actual stress levels vary by region, direction of

mining, and surface topography.

Overwide intersections: It is suspected that in weak rock,

small increases in intersection span (3 to 8 ft) can significantly

weaken the roof. Many roof falls in the study were either

inaccessible or not cleaned up, making it difficult to document

accurately any overspans that may have contributed to the fall.

Quality of roof bolt installations: One of the most difficult

parameters to measure is the quality of roof bolt installation. It

is suspected that some failures can be caused by deficient

bolting practices, including loss of tension, large bolt annulus,

bent or notched bolts, overdrilled holes, or long lag times before

bolting. All of these factors may mask the performance of bolt

systems by increasing the roof fall rate.

OTHER RESULTS

As described above, the statistical analysis of data becomes

more complicated with increasing numbers of variables.

Interdependence of variables and errors in measurement are

compounded with large numbers of variables. An alternative

was to conduct analyses using "paired data" from individual

mines. In these cases, only a single variable changes.

The most successful of these analyses was on roof bolt

length. From the large data set, 13 pairs of data where

two different lengths of roof bolts were used at the same mine

were extracted. The roof bolt lengths differed by at least 1 ft in

the pairs. Table 3 shows the bolt lengths, along with the

CMRR, the bolt type, the roof fall rate, and the percentage of

difference in roof fall rate between the two lengths.

The data show that in 11 of 13 cases, the four-way intersection roof fall rate was less with the longer bolt (figure 20). The

roof fall rates for the paired data range from 0.0 to 18.3 and the

Table 3.–Test cases showing the effect of bolt length on roof fall rate

Mine CMRR Bolt length, ft Roof fall rate

(falls per 10,000 ft) % change Bolt type

1 . . . . . . . . . . 50 4 1.08 Fully grouted.

6 .66 -39 Fully grouted.

2 . . . . . . . . . . 37 6 12.07 Tension.

8 8.62 -29 Tension.

3 . . . . . . . . . . 41 5 1.28 Fully grouted.

6 .23 -82 Fully grouted.

4 . . . . . . . . . . 55 4 4.05 Fully grouted.

6 .79 -80 Fully grouted.

5 . . . . . . . . . . 58 3.5 .88 Fully grouted.

5 .23 -74 Fully grouted.

6 . . . . . . . . . . 39 5 1.79 Fully grouted.

6 .36 -80 Fully grouted.

7 . . . . . . . . . . 42 4 2.9 Tension.

6 1.11 -61 Fully grouted.

8 . . . . . . . . . . 42 4 18.3 Tension.

6 .76 -96 Fully grouted.

9 . . . . . . . . . . 44 4 0 Tension.

6 3.57 +100 Tension.

10 . . . . . . . . . 40 5 .52 Fully grouted.

8 0 -100 Tension.

11 . . . . . . . . . 40 5 .39 Fully grouted.

6 .26 -34 Tension.

12 . . . . . . . . . 30 5 2.8 Tension.

6 4.0 +40 Fully grouted.

13 . . . . . . . . . 50 4 2.45 Tension.

5 1.77 -28 Fully grouted.
69

 Figure 20.—Paired cases of long and short bolts showing benefits of long bolts and

decreasing roof fall rates.

decrease in roof fall rate (average 65%, n ' 13) with increasing

bolt length holds true even in the high roof fall rate range. Six

of the 13 pairs of bolt lengths mixed tensioned bolts with fully

grouted bolts and the relationship holds true with both types of

bolts. Thus, through a wide range of CMRR (30-58), an

increase of at least 1 ft in bolt length can be expected to result

in a decrease in four-way intersection roof fall rate.

The paired bolt length data contain four cases comparing

longer fully grouted bolts with shorter tensioned bolts. The roof

fall rate was lower in just one case and higher in the three other

cases when the mine used the shorter tensioned bolts.

The relationship between CMRR and intersection span was

also analyzed. The data were partitioned by four-way intersection fall rate into low (0-0.001 falls per 100 four-way

intersections), moderate fall rate (0.001-0.05 falls per 100 fourway intersections), and high fall rate (>.05 falls per 100 fourway intersections). Additionally, only fully grouted bolts were

used in the analysis. By logistic regression, a line was fitted to

the data and presented in figure 21. No cases with high roof fall

rates fell to the right side of the regression line. This line can

be used to indicate whether smaller spans might be helpful in

relieving the incidence of roof falls. It is rare for high roof fall

rates to occur in roof with intersection spans less than the

equation. The less conservative regression line (span ' 31 %

0.66 CMRR) might be an appropriate first approximation

design equation. Based on our data, there is also a likelihood

that intersection falls will be reduced by a decrease in

intersection span. The intersection span measured for the study

was taken at the midpoint between the original rib corner and

the subsequent sloughage point. This differs from MSHA’s

measurement point, which is the original rib corner. For this

reason, the projected intersection spans will be somewhat

conservative.

Intersection diagonals are usually related to entry width. The

data were studied to determine the typical intersection spans

that are encountered underground. Figure 22 shows the mean

of the sum of the diagonals for 16-, 18-, and 20-ft entries. It

also shows that in deeper mines, the sum of the diagonals was

3-4 ft wider than in the shallow mines with the same entry

width, probably because of greater rib sloughage.
70

Figure 21.—Relationship between CMRR, intersection span, and roof fall rate.

Figure 22.—The effect of overburden on intersection diagonal.
71

CONCLUSIONS

Data on roof quality and roof bolt performance were

collected from interviews and underground reconnaissance at

41 U.S. coal mines. The roof fall rate (falls per 10,000 ft of

drivage) and the four-way intersection fall rate (roof falls in

four-ways/total number of four-ways) were developed as the

outcome variables for analyzing the influence of numerous

geotechnical variables on stability. From the data it was

determined that higher roof fall rates were more common in the

lower CMRR range (CMRR # 50). Intersections were much

more likely to fall than entry segments, and four-way intersections were more likely to fall than three-way intersections.

When the data were divided into two groups by depth of cover,

a relationship between PRSUP and the CMRR was determined

that could be used in design. The study determined that

overburden depth could be used as a surrogate for stress level.

Paired data extracted from the database show that increasing

bolt length decreased the roof fall rate in 11 of 13 cases over a

wide range of roof fall rate and bolt types. A useful

relationship between the intersection span and the CMRR was

also found. The data showed considerable scatter, which was

attributed to variations in roof geology, horizontal stress, and

bolt installation quality, none of which could be measured.

The method for constructing historic roof bolt maps and

hazard maps using the CMRR was described. These methods

of tracking roof quality and support performance will be

valuable for individual mine operators.

REFERENCES

Chase FE, Mark C, Mucho TP, Campbell PL, Holbrook AD [1999]. The

advance and relieve mining method: a horizontal stress control technique. In:

Peng SS, ed. Proceedings of the 18th International Conference on Ground

Control in Mining. Morgantown, WV: University of West Virginia, pp. 300-

308.

Dolinar DR, Bhatt SK [2000]. Trends in roof bolt applications. In: New

Technology for Coal Mine Roof Support. Pittsburgh, PA: U.S. Department of

Health and Human Services, Public Health Service, Centers for Disease

Control and Prevention, National Institute for Occupational Safety and Health,

DHHS (NIOSH) Publication No. 2000-151, IC 9453.

Dolinar DR, Oyler DC, Compton CS [1997]. Analysis of extensometer

data from a room widening experiment designed to induce a roof fall. In:

Peng SS, ed. Proceedings of the 16th International Conference on Ground

Control in Mining. Morgantown, WV: University of West Virginia, pp. 289-

295.

Mark C [1999a]. Application of the Coal Mine Roof Rating (CMRR) to

extended cuts. Min Eng 51(4):52-56.

Mark C [1999b]. Empirical methods for coal pillar design. Proceedings

of the Second International Workshop on Coal Pillar Mechanics and Design.

Pittsburgh, PA: U.S. Department of Health and Human Services, Public Health

Service, Centers for Disease Control and Prevention, National Institute for

Occupational Safety and Health, DHHS (NIOSH) Publication No. 99-114,

IC 9448, pp. 145-154.

Mark C [2000]. Design of roof bolt systems. In: New Technology for Coal

Mine Roof Support. Pittsburgh, PA: U.S. Department of Health and Human

Services, Public Health Service, Centers for Disease Control and Prevention,

National Institute for Occupational Safety and Health, DHHS (NIOSH)

Publication No. 2000-151, IC 9453.

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stress. In: New Technology for Longwall Ground Control; Proceedings: U.S.

Bureau of Mines Technology Transfer Seminar. Pittsburgh, PA: U.S.

Department of the Interior, U.S. Bureau of Mines, SP 01-94, pp. 53-76.

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systems using roof classification. In: New Technology for Longwall Ground

Control; Proceedings: U.S. Bureau of Mines Technology Transfer Seminar.

Pittsburgh, PA: U.S. Department of the Interior, U.S. Bureau of Mines,

SP 01-94, pp. 5-18.

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a practical rock mass classification for coal mines. Pittsburgh, PA: U.S.

Department of the Interior, Bureau of Mines, IC 9387.

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performance in high stress conditions. In: Peng SS, ed. Proceedings of the

14th International Conference on Ground Control in Mining. Morgantown,

WV: West Virginia University, pp. 30-38.

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design better mine openings. Eng Min Dec:76-80.

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model. In: Proceedings of the 30th U.S. Symposium on Rock Mechanics.

Morgantown, WV: West Virginia University, pp. 503-510.

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Conference on Ground Control in Mining. Morgantown, WV: University of

West Virginia, pp. 249-256.

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Mining. Golden, CO: Colorado School of Mines, pp. 179-191.
73

LOAD BEHAVIOR OF GROUTED BOLTS IN SEDIMENTARY ROCK

By Stephen P. Signer1

ABSTRACT

This paper presents an overview of laboratory and field tests on approximately 250 fully grouted roof bolts

instrumented with strain gauges in order to study loading behavior. Laboratory work included pull tests, timedependent tests, and shear tests. The field tests were conducted in 14 different mines extracting different

commodities. However, because of the focus of this publication, only the results of tests in 9 coal mines are

reported here. In the field tests, all but 14 bolts were loaded by rock movement; these 14 bolts were tested

using pull gear. The variables studied included anchorage length, mechanical interlock, time-dependent

behavior, shear, installation load, loading rate, and maximum load level. Such information will improve the

understanding of the behavior of resin-grouted bolts, which in return will enhance the safety of miners by

reducing the occurrence of rock falls. The influences of geology, overburden depth, entry width, mininginduced stresses, and bolt spacing are not described.

1

 Mining engineer, Spokane Research Laboratory, National Institute for Occupational Safety and Health, Spokane, WA.
74

INTRODUCTION

Millions of roof bolts are installed in U.S. coal mines each

year to prevent ground falls. Despite the importance of entry

stability, there is a lack of understanding of how roof bolts

provide reinforcement to the mine rock. Bolt loads are affected

by many variables, such as changes in mining conditions

(geology, geometry, and in situ stress fields) and bolt properties

(diameter, length, spacing, and stiffness). This lack of understanding has contributed to our inability to prevent rock falls,

which remains one of the most significant safety hazards in

underground mines. The Spokane Research Laboratory (SRL)

of the National Institute for Occupational Safety and Health

(NIOSH) has conducted research to study the behavior of

grouted roof bolts and how they reduce support failure and

subsequent ground falls. The purpose of this paper is to summarize important facts and principles obtained from the

completed studies. Because of space limitations, we will refer

to previously published material that contains detailed information on each individual test setup and variables.

The first investigation [Serbousek and Signer 1987] was a

study of the axial elastic behavior of grouted bolts installed in

concrete blocks. The purpose was to find out how load was

transferred both between the bolt and the grout and the grout

and the rock by using a known loading condition uninfluenced

by geologic variations. Strain gauges were installed to measure

load changes along the length of the bolt. More than 50 pull

tests were performed in which applied loads were restricted to

the elastic range of the bolt steel. Variations were made in hole

size, bolt length, grout type, and grout strength. Results of an

axisymmetric, finite-element numerical model were compared

with these test results.

The second phase of research [Signer 1990] was performed

at four different coal mines in Colorado, Illinois, and

Pennsylvania. Fourteen instrumented bolts were installed in

sedimentary rock and tested using the same procedures as

employed in the laboratory tests. In addition, these bolts were

loaded to failure to study their nonlinear behavior. The purpose

of this investigation was to verify the results of the axial elastic

laboratory studies in the field where geology became a variable.

A second goal was to study load transfer mechanisms when

bolts were loaded past their elastic limit.

A third study [Signer 1988] involved laboratory tests in concrete blocks to study the creep behavior of grouted bolts instrumented with strain gauges. Variations were made in bolt

length and grout type. The loads were applied with pull gear

and held constant for long periods of time. The purpose of this

investigation was to see if a constant load applied to the bolt

head would propagate along the length of the bolt over time.

Laboratory tests [McHugh and Signer 1999] were also conducted to study the axial and bending behavior of grouted bolts

subjected to horizontal joint movements. These tests were conducted on 17 instrumented bolts installed perpendicular to a

joint surface in 0.6- by 0.6-m (2- by 2-ft) square concrete

blocks. A shear force was applied until the bolts failed.

The last phase of research consisted of field studies in which

fully grouted bolts instrumented with strain gauges were installed as supports and loaded by mining-induced stress changes

[Signer and Jones 1990; Signer et al. 1993; Maleki et al. 1994;

Larson et al. 1995; Signer and Lewis 1998]. These bolts were

placed in 14 mines (9 coal, 2 trona, 2 gold, and 1 platinum).

Variations in bolt length, bolt spacing, bolt diameter, geology,

geometry, stress fields, and mining methods were recorded to

study the response of grouted bolts under actual field conditions. These investigations have just been completed, and a

complete data analysis is ongoing.

BEHAVIOR OF GROUTED ROOF BOLTS

BOLT LOADING

Roof bolts are loaded in four ways: axial, bending, shear,

and torsion (figure 1). Torsional loading applies only to bolts

that are actively tensioned by rotation and are not covered in

this paper. Steel bolts are weakest in shear and strongest in

axial. Although axial bolt loads result from by both vertical and

horizontal rock movement, most of the axial load is produced

by vertical rock movement. Because there is a large difference

in strength between the steel and the rock, actual shear failures

of the steel are not common in sedimentary deposits. If a shear

plane develops in the bolted rock mass and horizontal

movement occurs, then the bolt will be subjected to a combination of axial and bending loads. A commonly observed phenomenon is the S-shaped failure shown in figure 2. This characteristic is actually a combination of axial and bending forces.

VERTICAL ROCK MOVEMENT

A grouted bolt is a passive support system, which means that

bolt loads are caused by rock movements. Additional rock

movement increases load on the bolt and reduces and/or stabilizes movement in the mine rock. Vertical rock movements

produce axial loading in the bolt. Axial loads are transferred

between the bolt and the rock by shear resistance in the grout.

This resistance is the result of mechanical interlock in which

load is transferred between the steel bolt and the grout and the

grout and the rock via contact surfaces. Bolt hole walls have

voids and irregularities created by the drilling process. Steel

bolts are rolled with ribs to provide anchorage. Grout fills these

irregularities and voids in the walls and bolts if the bolt has

been properly installed. Localized deformation and crushing

will occur in the grout at the contact points before the system
75

Figure 1.—Types of roof bolt loads.

Figure 2.—Bolt response to

shear movement.

tightens up, which allows additional deflection in both the bolt

and the rock. Mechanical interlock will cause shear forces to be

transferred from one medium to another until the maximum

shear strength of the grout and/or rock is reached. At that point,

the weakest material will fail, and friction will control load

transfer. The rate of load transfer is similar to an exponential

decay curve and is dependent on the material properties of the

bolt, the grout, the rock, and the respective interfaces.

The amount of bolt length necessary to transfer the ultimate

axial load capacity of the bolt to the rock is called anchorage

length or critical embedment length and is a function of the

strength of both the rock and the grout. If anchorage length is

not long enough, then the grout or rock interfaces will fail, and

the bolt will pull out of the hole.

HORIZONTAL ROCK MOVEMENT

Horizontal movement of a rock joint will compress the

surrounding rock against the grout. This produces shear loads in

the bolt. If the forces exceed the compressive strength of the

rock, the rock will fracture. When this occurs, shear loads will be

reduced and will be replaced by a combination of both bending

and axial loads. Bending loads are highest near the joint and will

dissipate quickly along the length of the bolt. Axial loads will

travel farther along the length of the bolt than bending loads and

will interact with the grout and rock as noted above.

FAILURE

Various types of failure can occur when using grouted bolts.

Failure can take place in the bolt, the grout, the rock, or at the

bolt-grout or grout-rock interfaces. The type of failure depends

on the characteristics of the system and the material properties

of individual elements.

(1) If the rock is weaker than the grout and if anchorage

length is inadequate, then failure of the bolt system will occur

at the grout-rock interface, which is the weakest point. As the

shear strength of this interface is exceeded, then failure will

progress from the point of maximum load in the bolt down the

length of the bolt.

(2) If the grout is weaker than the rock, then shear failure

will occur in the grout at the bolt-grout interface. If the anchorage length is inadequate, then failure will progress along

the length of the bolt. Shear stress is greater at the bolt-grout

interface than it is at the grout-rock interface simply because

there is less area at the bolt-grout interface. The bolt-grout interface is also more prone to failure because of Poisson's effect

on the steel bolt, which causes the bolt to pull away from the

grout as the bolt is loaded.

(3) If there is adequate anchorage length to develop the full

capacity of the steel bolt, regardless of the properties of the

grout and the bolt, then the bolt will fail if loading on the bolt

exceeds the ultimate strength of the steel. However, prior to

bolt failure, localized grout or rock shear failure will occur.

This is because the bolt, which has a greater ductility, will take

larger deflections than the rock or the grout.
76

Figure 3.—Load transfer of No. 6 bolts in concrete blocks.

Figure 4.—A smooth hole lowers anchorage

strength, as seen with this bolt as it pulls out of

the hole.

RESULTS

Approximately 250 bolts instrumented with strain gauges

were installed in both the laboratory and the field to study

loading behavior on fully grouted roof bolts. Laboratory tests

included pull tests, time-dependent tests, and shear tests.

Although the field tests were conducted in 14 different mines

extracting different commodities, because of the focus of this

publication, only the results of tests in 9 coal mines are reported

here. In the field tests, all but 14 bolts were loaded by rock

movement; these 14 bolts were tested using pull gear. The

influences of geology, overburden depth, entry width, mininginduced stresses, and bolt spacing are not described. The

following discussion covers some of the major factors related

to the performance of fully grouted bolts.

BOLTS LOADED UNDER ARTIFICIAL CONTROL

Anchorage Length and Mechanical Interlock

Elastic tests in which grout type, hole size, and bolt length

were varied were conducted in the laboratory on 50 bolts

[Serbousek and Signer 1987]. Results on the 1.2- and 0.6-m (4-

and 2-ft) long bolts indicated that 56 cm (22 in) of bolt length

were required to transfer 90% of the load from the bolt to the

rock (figure 3). Polyester resin and gypsum grout were used

with a 19-mm (0.75-in) bolt and installed in 25-mm (1-in) holes.

Nineteen-millimeter (0.75-in) bolts were also tested in 35-mm

(1-3/8-in) holes using gypsum grout. The variations in grout

type and hole size had no statistically significant effect.

The results from the axial elastic test conducted on grouted

bolts installed in shale compared well with the results from

previous laboratory work. The average anchorage length for

these bolts was slightly longer than for bolts installed in concrete

blocks, even though the field test results showed more variability.

The roof at the first mine site contained layers of weaker rock.

Test results from this mine reflected the presence of these weaker

layers as changes in the rate of load transfer. A weaker layer requires a longer anchorage length compared to that needed in

stronger rock. The stiffness of the bolting system decreases in

weaker zones due to slip at the grout-rock interface.

Laboratory and field studies of fully grouted bolts indicate

that the average anchorage length for bolts in competent rock is

56 cm (22 in). Anchorage length in weak and broken rock must

be determined by field pull tests of bolts with grout lengths less

than 30 cm (1 ft). Anchorage length can also be affected if the

bolt hole is smooth, which reduces the effect of mechanical

interlock, so that the resisting force is mostly friction. The effect of a pull test on a 30-cm (1-ft) long bolt in a smooth hole is

shown in figure 4.

Anchorage length depends on the material properties of the

bolt, the grout, and the rock; the quality of the installation; the

smoothness of the drill hole; and possibly other factors. Weaker grout and/or rock will require longer bolt anchorage lengths.

Proper installation of the bolt is critical to the performance of

the bolt. If the grout is inadequately mixed, is overspun, or is

glove-fingered, then the capacity of the grout to provide

mechanical interlock is severely impaired. Glove-fingering occurs when the plastic casing of the resin remains intact and

causes a plastic interface between the grout and the rock. The

bolt hole must be drilled with bits of appropriate sizes to

produce holes of the proper diameter [Pettibone 1987].

Readings from seven bolts were averaged, and the results are

shown in figure 5. Each curve represents load decay along the

bolt length. The curve was established from readings of the

applied load to the bolt and strain gauges. The length necessary
77

Figure 5.—Results of pull tests on instrumented bolts.

to transfer all the load from the bolt to the rock was the same at

different load levels. The slope of each curve is an indication

of the stiffness of the system. Increasing the applied load resulted in higher stiffnesses, but the load transfer length remained the same, indicating that mechanical interlock between

the bolt, the grout, and the rock was the primary mechanism for

transference of load. If adhesion were the mechanism of load

transfer, then stiffness would be the same for all elastic loads

and the anchorage length would increase as a function of

applied load. Friction could not be the load transfer mechanism

because bolt deflections were elastic.

Fifteen instrumented bolts were tested past the yield point of

the steel to study load transfer mechanisms prior to bolt failure.

The results show that yielding of the steel will translate down

the length of the bolt from 23 to 51 cm (9 to 20 in), depending

on rock strength, bolt hole properties, and bolt installation

quality. When the bolt has yielded, load is estimated using a

load strain curve that was determined experimentally. These

bolts were pulled an average of 5 cm (2 in) before failure.

Time-Dependent Behavior

Laboratory tests were conducted on six bolts to determine

the time-dependent properties of grouted bolts. These bolts

were instrumented with strain gauges and installed in 2.5-cm (1-

in) holes with both gypsum and resin grout. Bolts that were 1.2,

0.6, and 0.3 m (4, 2, and 1 ft) long were tested at applied loads

of 40, 58, 80, and 102 kN (9,000, 13,000, 18,000, and

23,000 lb). The strain gauges were monitored constantly to detect load changes along the bolt. Load was applied with hydraulic rams and maintained with hydraulic accumulators for a

period of at least 1 month, or until the bolt stabilized.

Figure 6 shows the results for the 1.2-m (4-ft) long bolts. To

determine the rate of load change per day, a linear fit was done

at each strain gauge location beginning at day 10 and continuing

to the end of the test. The resulting values were normalized by

dividing the rate of load change per day by the load on the bolt

at a given location. When the load on the bolt approached zero,

the data were deleted because of problems caused by dividing

by zero. Bolts installed with gypsum grout showed three to five

times more creep than bolts installed with resin grout.

Shear Tests

Laboratory tests on 17 bolts were conducted to study the

behavior of roof bolts subjected to shear loading over a range

of axial bolt loads. Fourteen strain gauges were attached to

each bolt to measure both axial and bending loads. The instrumented bolts were grouted through two high-strength concrete

blocks, and axial tension up to 75% of the yield strength was

applied to the bolts. The block interface was smooth and acted

as the failure plane. Shear loads were applied to the blocks

until the bolts failed due to shear movement of the joint. The

tests characterized the relationship of axial and bending loads

on the bolt to shear forces across a rock bedding plane.

The results showed that (1) axial bolt loading had little to no

effect on the capacity to resist shear loading of a joint, (2) the

bolts failed as a combination of axial and bending loads rather

than shear loads, (3) the bending loads caused by joint movement dissipated within a few inches of the joint, but axial loads

were transferred along the bolt length for the distance of the

anchorage length, and (4) all but one bolt failed within 5 cm

(2 in) of horizontal movement of the joint.

BOLTS LOADED BY ROCK MOVEMENT

Installation Loads

When a fully grouted bolt is installed, the upward thrust of the

roof bolter will compress the weakest rock layer in the immediate

roof. The amount of compression will depend on the distance

from the roof line, the strength of the surrounding rock layers, and

the amount of upward thrust of the roof bolter. The bolt will be

held in place until the grout hardens. When the force from the

roof bolter is removed, the compressed roof rock will rebound,

which produces a resisting force in the grouted bolt.

Installation loads on 40 strain-gauged roof bolts from

4 different coal mines are shown in figure 7. The data represent

an average of 420 strain gauges. Some bolts had 10 strain

gauges per bolt, and some bolts had 12 strain gauges per bolt.

The average installation load was 11 kN (2,400 lb) with a

standard deviation of 8.5 kN (1,900 lb). This means that installation loads tended to vary from zero to 20 kN (4,500 lb).

The highest initial load was usually near the roof line and

decreased as the distance from the roof line increased. The

amount of force developed depended on the properties of the

immediate roof, the ability to apply the upward thrust of the

roof bolter, and the behavior of the roof bolter operator. That

is, often a bolter operator would not apply the same amount of

thrust to the specially instrumented roof bolts as to a normal

bolt. The amount of time elapsed from installation to the first

measurement could affect the readings if the mine roof were

active. If the hole was not drilled deep enough, the bolt could

bottom out.
78

Figure 7.—Installation loads on instrumented bolts.

Figure 6.—Time-dependent behavior of (A) gypsumgrouted bolts and (B) resin-grouted bolts.

Figure 8.—Typical bolt load response to mining-induced

stresses.

Loading Profiles

Figure 8 shows axial loading behavior of a typical fully

grouted roof bolt. When the bolt is first installed, there is an

initial load that varies along the length of the bolt. As mining

progresses, redistribution of rock stresses change the stress

pattern in the immediate mine roof. A fully grouted bolt

responds to the stress changes by increasing in load. The

distribution of load along the length of the bolt varies in

proportion to the stress pattern. After primary mining has

reached a sufficient distance away from the test area, the load

on the bolt increases at a slower rate, stabilizes, and in some

cases decreases. If secondary mining occurs and the immediate

roof is within the zone of influence from abutment stress

changes, then the bolts will go through the same cycle of load

increase and stabilization. This sequence assumes that there are

enough bolts installed to cause the immediate roof to stabilize.

Loading Rates

The rate at which grouted bolts increase in load depends on

the rate of rock movement in the immediate roof. Stress

changes in the rock result in movement. The amount of movement depends on the rock and joint modulus and strengths.

Strain gauges on bolts can measure rock movement several

orders of magnitude finer than can other deflection measurement systems. For example, 0.05 mm (0.002 in) of movement in a bolt section 2.5 cm (1 in) long will result in 2,000

microstrain, which is the yield strain for a No. 6 slotted bolt.

Initial loading rates on 62 strain-gauged roof bolts installed

in 6 different coal mines are shown in figure 9. The initial loading rate was calculated by a linear fit of the change in axial

loads right after installation. The duration of this loading rate

varied significantly from one site to another, as did the load

level attained during initial loading.

The average rate of change of bolt load resulting from entry

development varied from 1 to 16 kN/d (250 to 3,500 lb/d). Each

bar on the graph represents an average of all the strain gauges at

each mine site. Each site had significant variations in loading

rates among bolts and even among gauges on each bolt. The

maximum loading rate at each site is significant because highly
79

Figure 9.—Initial load rate and rate of bolt load increase

after installation.

Figure 11.—Maximum axial loads on bolts in all coal mines.

Figure 10.—Strain-load curve for No. 6 and No. 7 slotted

bolts.

Figure 12.—Maximum axial loads on (A) No. 6 bolts and

(B) No. 7 bolts.

Table 1.—Properties of bolts

Bolt type Yield load Ultimate load

Crosssectional

area

kN lb kN lb cm2 in2

No. 6 . . . . . . . . 107 24,000 176 39,500 2.58 0.40

No. 6, slotted . . 96 21,500 160 36,000 2.38 0.37

No. 7 . . . . . . . . 160 36,000 270 60,500 3.87 0.60

No. 7, slotted . . 149 33,500 249 55,900 3.61 0.56
80

loaded bolts fail first. The maximum loading rate after bolt installation varied from 4.4 to 105 kN/d (1,000 to 23,500 lb/d).

There are many factors that cause these variations, including

geology, seam height, entry width, pillar size, and stress field.

MAXIMUM BOLT LOADS

Figure 10 shows the load-strain behavior of slotted bolts, and

table 1 compares the yield and ultimate strengths of slotted bolts

and regular rebar bolts. Bolt row spacing for the instrumented bolts

was reduced to compensate for the 10% reduction in strength. In

some cases, the instrumented bolts were installed as supplementary

support. Also, No. 7 strain-gauged bolts were sometimes installed

where No. 6 bolts were used as the support system.

The distribution of maximum axial loads, i.e., the maximum

load on each instrumented bolt, is shown for all bolts installed

in coal mines in figure 11 and for No. 6 bolts and No. 7 bolts in

figure 12A and 12B, respectively. This value, rather than average load, is what would cause a bolt to break. Maximum

axial load on the majority of the bolts exceeded the yield point

of the steel, but was less than ultimate load. Maximum load on

the instrumented bolts can be used as an estimate for selecting

bolt size and spacings.

These data represent loading on 92 instrumented bolts at

eight different mine sites. Seventy-five percent of the instrumented bolts reached the yield point of the steel on at least one

gauge location, and 50% of the bolts exceeded the yield point

of the steel. These values represent bolt loading resulting from

a variety of loading conditions. Most were installed in longwall

gate road entries; loading represents both passes of the longwall. Fourteen bolts were installed in a longwall recovery room

where the roof support failed.

CONCLUSIONS

The anchorage length for grouted bolts installed in competent rock is 56 cm (22 in) and is established by mechanical

interlock of the grout. Bolts loaded to failure show that yield

will translate down the length of the bolt from 23 to 51 cm (9 to

20 in) and will be deflected an average of 5 cm (2 in) before

failure. Load creep on bolts installed with polyester resin grout

was shown to be minimal compared with creep on bolts installed with gypsum grout. Results of shear tests showed that

the bolts failed from a combination of axial and bending loads

rather than shear loads and that axial bolt loading had little

effect on the capacity to resist shear loading of a joint.

The average load on fully grouted bolts just after installation

was 11 kN (2,400 lb) and increased anywhere from 1 to 105

kN/d (250 to 23,500 lb/d). The data representing 92 instrumented bolts at 8 different coal mine sites showed that 75% of

the bolts reached the yield point of the steel (0.2% strain) and

50% of the bolts exceeded the yield point of the steel. 

The results of this study will increase understanding of the

behavior of roof bolts in underground mines, which in turn will

enable miners to select the appropriate bolts for reducing roof

falls, thereby increasing workplace safety.

REFERENCES

Larson MK, Stewart CL, Stevenson MA, King ME, Signer SP [1995]. A

case study of a deformation mechanism around a two-entry gate-road system

involving probable time-dependent behavior. In, Peng SS, ed. Proceedings of

the 14th International Conference on Ground Control in Mining. Morgantown,

WV: West Virginia University, pp. 295-304.

Maleki H, Signer SP, King ME, Edminster PA [1994]. Evaluation of

support performance in a highly stressed mine. In: Peng SS, ed. Proceedings

of the 13th International Conference on Ground Control in Mining.

Morgantown, WV: West Virginia University, pp. 9-17.

McHugh E, Signer SP [1999]. Roof bolt response to shear stress:

laboratory analysis. In: Peng SS, ed. Proceedings of the 18th International

Conference on Ground Control in Mining. Morgantown, WV: West Virginia

University, pp. 232-238.

Pettibone HC [1987]. Avoiding anchorage problems with resin-grouted

roof bolts. Spokane, WA: U.S. Department of the Interior, Bureau of Mines,

RI 9129.

Serbousek MO, Signer SP [1987]. Linear load transfer of fully grouted

roof bolts. Spokane, WA: U.S. Department of the Interior, Bureau of Mines,

RI 9135.

Signer SP [1988]. Comparative studies in the mechanics of grouted roof

bolts. In: Peng SS, ed. Proceedings of the Seventh International Conference

on Ground Control in Mining. Morgantown, WV: West Virginia University,

pp. 282-288.

Signer SP [1990]. Field verification of load transfer mechanics of fully

grouted roof bolts. Spokane, WA: U.S. Department of the Interior, Bureau of

Mines, RI 9301.

Signer SP, Jones SD [1990]. A case study of grouted roof bolt loading in

a two-entry gate road. In: Peng SS, ed. Proceedings of the Ninth International

Conference on Ground Control in Mining. Morgantown, WV: West Virginia

University, pp. 35-41.

Signer SP, Lewis JL [1998]. A case study of bolt performance in a twoentry gateroad. In: Peng SS, ed. Proceedings of the 17th International

Conference on Ground Control in Mining. Morgantown, WV: West Virginia

University, pp. 249-256.

Signer SP, Mark C, Franklin G, Hendon G [1993]. In: Peng SS, ed.

Comparisons of active versus passive bolts in a bedded mine roof. In:

Peng SS, ed. Proceedings of the 12th Conference on Ground Control in

Mining. Morgantown, WV: West Virginia University, pp. 16-23.
81

SUMMARY OF FIELD MEASUREMENTS OF ROOF BOLT PERFORMANCE

 By Christopher Mark, Ph.D.,1

 Dennis Dolinar,2 and Thomas P. Mucho3

ABSTRACT

During the 1990s, the former U.S. Bureau of Mines conducted a number of field studies in which the

performance of different types of roof bolts were evaluated in different geologic environments. The studies

used a standard suite of measurements, including multipoint extensometers, strain-gauged roof bolts, and roof

bolt load cells. The sites were chosen to investigate the effect of a variety of parameters, including installed

tension, bolt capacity, grout annulus, and horizontal stress orientation. Although not fully successful, the

measurements provided valuable insights into each of these issues. They also showed that instrumentation and

monitoring have important advantages over observational methods for comparing the performance of different

roof bolting systems.

1

Supervisory physical scientist.

2

Mining engineer.

3

Supervisory physical scientist.

Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
82

INTRODUCTION

Roof bolts interact with the ground to create a reinforced

rock structure. The mechanics of this interaction are difficult,

if not impossible, to replicate in the laboratory. Field studies

are essential to developing an understanding of how factors

such as bolt tension, bolt length, bolt capacity, and resin

annulus contribute to the support of real rock masses. Detailed

measurements of bolt loads and roof movements can provide

the information necessary to build conceptual and numerical

models of supported mine roof.

Field studies are also the only way to compare the overall

effectiveness of different roof bolt systems. In the United

States, such comparisons are usually made by visual observations rather than by measurements. If an area supported

by one type of bolt experiences less roof degradation or fewer

roof falls than an area supported by another type, then the first

bolt is deemed superior [Stankus 1991]. This observational

approach, however, has limitations. Often, significant roof

movements can occur without visual evidence at the roof line.

Waiting to see how many roof falls occur can be expensive,

particularly if large areas of the mine were supported with a

particular bolt before its inadequacies became apparent. Again,

instrumentation can provide an alternative. Measurements can

show that bolts are overloaded or that the roof is becoming

unstable long before there is any visual evidence.

Studies of roof bolt behavior have a long history in the

United States. Some of the classic early work with straingauged resin bolts was performed by Karabin and Debevec

[1976] and Haas [1981], followed by Serbousek and Signer

[1987] and Signer [1990]. Other insights regarding the

interaction between roof bolts and the rock mass came from

researchers in Australia [Gale 1991; Hurt 1992].

In the early 1990s, the former U.S. Bureau of Mines

(USBM) embarked on a major program of roof bolt field

studies. One group of studies focused on the behavior of fully

grouted, nontensioned resin bolts and is reported by Signer

[2000]. The second group of studies, which is described here,

had two main goals:

1. To study fundamental aspects of roof bolt performance

by comparing different types of bolts in a variety of geologic

environments; and

2. To develop an effective instrumentation plan for evaluating roof bolt systems at a particular site.

Ultimately, studies were conducted at 12 sites in 7 mines

(mines A through G; see table 1). Most of the studies were

conducted under cost-sharing Memorandums of Agreement

between the USBM and cooperating coal companies.

Unfortunately, the program ended in 1995 when the closing of

the USBM resulted in reduced funding for ground control

research.

Table 1.—Summary of field studies
83

Figure 1.—Standard instrumentation for field evaluation of roof bolt performance.

STANDARD INSTRUMENTATION

The studies usually began with an assessment of roof

geology using the Coal Mine Roof Rating (CMRR). The

CMRR rates the structural competence of the roof and allows

the roof at different sites to be compared on a single scale

[Molinda and Mark 1994]. Instrumentation sites also usually

included a 1-in (25-mm) diameter, 16-20 ft (5-6 m) vertical

borehole for logging with a stratascope. Stratascope surveys

provide a means to assess interfaces, bedding, or other

geological features; identify general roof rock lithology; and

observe bed separations.

Several means were used to monitor support loads in the

studies. For fully grouted resin bolts, strain gauges were

mounted in slots machined in the roof bolts [Signer 2000]. For

resin-assisted point-anchor systems, loads were monitored

using electronic or hydraulic load cells mounted between the

plate and the bolt head. Further details may be found in the

original reports [Signer et al. 1993; Mucho et al. 1995; Campoli

et al. 1996; Mark et al. 1998].

Roof movements were monitored using extensometers,

convergence stations, and observation holes. One to three

extensometers were included in each instrumented area

depending on the detail of roof movement required by the

investigation. Most of the studies used the Sonic Probe type of

extensometer. The Sonic Probe measures the distance between

magnetic anchors placed in a 38-mm (1.5-in) diameter borehole

to a claimed accuracy of 0.025 mm (0.001 in). As many as 20

anchors can be placed in a 6-m (20-ft) long vertical borehole.

The data from the extensometer locations can be presented

as deformation in each interval or as percent strain. Strain is

determined by dividing the movement between the anchor

intervals by the original length of the interval. Multiplying this

strain by 100 yields percent strain for the interval. A rule of

thumb developed abroad is that 1% strain measured above the

bolts is an unstable condition [Hurt 1992]. The concept is that

once a roof bed experiences 1% strain, it fails and can no longer

carry horizontal stress, thus forcing the stresses to move higher

into the roof. If a roof bed fails above the bolts, it may indicate

a loss of ground control. Large roof strains within the bolted

horizon are of much less concern. One goal of the studies was

to test whether the concept of roof strain was useful for U.S.

conditions.

In the course of the studies, a standard instrumentation plan

evolved (figure 1). Unfortunately (as the comments in table 1

indicate), it was not possible to install all of the instrumentation

at every site. No instrumented bolts were used at five of the

sites, and simple three-point extensometers were used at two

others.
84

DISCUSSION

Brief descriptions of the field sites and the individual results

are included in the appendix to this paper. Ultimately, the

number of sites was too small to provide definitive answers to

any of the fundamental questions about bolt mechanics.

Moreover, they do not allow a multivariate analysis, in which

interactions between the geologic and design factors might be

assessed. Nevertheless, the studies provided some valuable

insights into the effects of these factors on roof bolt

performance, which are discussed below.

GEOLOGY

The tests were performed in a wide variety of geologic

environments, with roof rocks that included coal, underclay,

shale, "stack rock" sandstone, and siltstone. The CMRR values

shown in table 1, however, indicate that the sites can be

classified into two groups. The CMRR was 40 or less at five of

the sites (mines B, D, and E); the roof there could be described

as "weak." At the remaining seven sites, the CMRR was

between 47 and 50, which means that the roof strength was

"moderate."

Roof strains in excess of 1% were measured above the bolt

anchorage at five sites, although no roof falls occurred. Three

of these were weak roof sites in the vicinity of longwalls,

indicating that weak roof is more likely to experience large roof

movements than moderate roof. Large strains were measured

within the bolted horizon at many sites in all types of roof,

however. No clear differences in bolt loading patterns emerged

among the different geologies.

HORIZONTAL STRESS

Horizontal stress was clearly a major factor at many of the

sites. At two mines (B and C), extensometers were placed in

entries and crosscuts that were oriented in "good" and "bad"

directions relative to the horizontal stress. In both cases, the

roof strains were at least three times greater in the "bad"

direction.

Several studies showed the effect of the "horizontal stress

abutment" due to longwall mining that has been described by

Mark et al. [1998]. The sites at mines A, D, and F were located

in longwall gate entries that were subjected to horizontal stress

abutments. Large roof movements were measured at each site,

and the majority of the instrumented roof bolts approached or

exceeded the yield point. In contrast, the sites at mine E were

located in a stress shadow, and no new roof movements

occurred as the longwall face approached. In fact, the bolt

loads even decreased slightly!

INSTALLED TENSION

One of the most controversial issues in roof bolting is the

im-portance of installed tension. Three studies compared

tensioned to nontensioned bolts, but the results were

ambiguous.

The study at mine C compared tensioned and nontensioned

fully grouted bolts. Greater movements were measured at the

nontensioned site, primarily within the bolted horizon. The

presence of a preexisting cutter in the roof at the nontensioned

site, however, may have influenced the results.

At mine G, resin-assisted point-anchor tensioned bolts were

compared with nontensioned fully grouted resin bolts. At a

third site, the fully grouted bolts were supplemented by some

resin-assisted point-anchor bolts. The point-anchor bolts were

1.5 m (5 ft long), while the fully grouted bolts were 1.8 m (6 ft)

long. Out of a total of seven instrumented intersections, two

experienced strains in excess of 1% above the bolts. One was

in the point-anchor site, but the other was one where the fully

grouted bolts had been supplemented by point anchors.

Overall, the differences among the three bolting systems were

probably not statistically significant.

The study at mine F was probably the most informative of

the three. Here, the resin-assisted point-anchor bolts clearly

provided better roof control than the nontensioned, fully

grouted bolts. The point-anchor bolts were also 60% stronger,

however, and their greater capacity may have accounted for

their better performance. More surprising were the bolt load

measurements. These showed that although the fully grouted

bolts were installed without tension, within days their loads

were equal to or greater than those of the point-anchor bolts.

The loadings on the two systems continued to increase at

approximately the same rate as mining progressed, up until the

point where most of the fully grouted bolts were loaded beyond

their yield point.

BOLT CAPACITY

As previously mentioned, the study at mine F found that the

higher capacity bolts performed better, although installed

tension may have been a contributing factor. However, mine F

subsequently switched to higher capacity, nontensioned, fully

grouted bolts and used them successfully for many years.
85

The study at mine D provided a clearer association between

greater capacity and better roof control. Here, the two types of

resin-assisted point-anchor tension bolts that were compared

were nearly identical except for the length of the resin used to

assist the mechanical shell. The bolts with the shorter length of

resin proved to have 40% less capacity due to inadequate

anchorage. The measurements showed that both types of bolts

loaded approximately in tandem until the short-anchor bolts

slipped. Then the loads on the short-anchor bolts diminished

while the roof movements accelerated. Ultimately, a 6% roof

strain was measured above the anchorage. In contrast, the loads

continued to increase on the long-anchor bolts, and roof control

was maintained.

The mine D study also demonstrated the value of instrumentation in evaluating different bolt systems. There was no

visible difference between the sites, but the measurements

clearly showed that the performance of the short-anchor bolts

was inadequate. Also, it was evident that measurements of both

roof movement and bolt load were necessary to tell the

complete story. Larger roof strains by themselves could have

meant that the roof was just more aggressive, while reduced

bolt loads alone might have signified more stable roof. But the

combination clearly signaled that the bolts had slipped and had

lost control of the ground.

BOLT LENGTH

Different lengths of bolts were studied at mines E and G.

No meaningful comparison was possible at mine E because all

of the sites were stable. At mine G, although the greatest roof

strains were measured where the shorter bolts were used, the

results were probably not statistically significant.

RESIN ANNULUS

The study at mine A was designed to investigate whether a

"reduced annulus" (a smaller difference between the bolt

diameter and the diameter of the bolt hole) would improve the

performance of the primary bolting system (see Mark [2000]

for a discussion of the importance of annulus to fully grouted

resin bolt performance). The instrumented bolts were installed

with annuluses of 7 mm (0.28 in) and 3 mm (0.12 in), but no

significant differences in the bolt-loading histories were

observed.

INTERSECTIONS VERSUS ENTRIES

Statistics clearly show that intersections, because of their

greater spans, are significantly more prone to roof falls than

entries. Measurements from the field studies provide further

confirmation. At mine D, roof strains in the intersections were

typically twice as great in the intersections, although the bolt

loadings were approximately the same in all locations. At mine

F, the bolt loads were typically 25% higher in the intersections,

while the measured roof sags were similar to those in the

entries.

BOLT LOCATION

The field studies found that, in general, the bolts with the

highest loads were located in the center of the entry or

intersection. When a cutter was present, such as at mines A, F,

and G, the bolts nearest the cutter were likely to be the most

heavily loaded.

CONCLUSIONS

The field studies were only partially successful in achieving

their goals. Of the bolt design parameters that were evaluated,

only bolt capacity seemed to clearly affect roof stability. The

results concerning installed tension, bolt length, and resin

annulus were all ambiguous.

On the other hand, the studies confirmed that greater roof

bolt loads and more severe roof movements are likely to occurC

$ In intersections;

$ Near roof cutters;

$ In entries perpendicular to the principal horizontal stress;

and

$ In areas subjected to horizontal stress concentrations.

It seems that in many cases roof stability could be improved

by selectively installing stronger and/or longer bolts in these

areas.

Finally, the standard instrumentation plan was shown to

be an effective approach to evaluating different roof bolting

systems. It provides an unequaled look into the performance

of the supports and their interactions with the roof. Hopefully, mines will continue to use the instrumentation to help

address their roof support issues and at the same time improve

our fundamental knowledge of how roof bolts work.
86

REFERENCES

Campoli AA, Rebuck D, Mucho TP, Molinda GM [1996]. State-of-the-art

room and pillar mining in the Kittanning Coalbed. In: Ozdemir L, Hanna K,

Haramy KY, Peng S, eds. Proceedings of the 15th International Conference

on Ground Control in Mining. Golden, CO: Colorado School of Mines,

pp. 269-285.

Gale WJ [1991]. Strata control utilizing rock reinforcement techniques and

stress control methods in Australian coal mines. The Mining Engineer (U.K.)

Jan:247-253.

Haas CJ [1981]. Analysis of rockbolting to prevent shear movement in

fractured ground. Min Eng June:691-704.

Hurt K [1992]. Rockbolting design in UK coal mines. In: Proceedings of

Eurock 92. London, U.K.: Thomas Telford, paper No. 71, pp. 414-419.

Karabin GJ, Debevec WL [1976]. Comparative evaluation of conventional

and resin bolting systems. U.S. Department of the Interior, Mine Enforcement

and Safety Administration, IR 1033.

Mark C [2000]. Design of roof bolt systems. In: New Technology for Coal

Mine Roof Support. Pittsburgh, PA: U.S. Department of Health and Human

Services, Public Health Service, Centers for Disease Control and Prevention,

National Institute for Occupational Safety and Health.

Mark C, Mucho TP, Dolinar DR [1998]. Horizontal stress and longwall

headgate ground control. Min Eng Jan:61-68.

Molinda GM, Mark C [1994]. The Coal Mine Roof Rating (CMRR):

a practical rock mass classification for coal mines. Pittsburgh, PA: U.S.

Department of the Interior, Bureau of Mines, IC 9387.

Mucho TP, Mark C, Zelanko JC, Compton CS [1995]. Roof support

performance in high stress conditions. In: Peng SS, ed. Proceedings of the

14th International Conference on Ground Control in Mining. Morgantown,

WV: West Virginia University, pp. 30-38.

Serbousek MO, Signer SP [1987]. Linear load-transfer mechanics of fully

grouted roof bolts. Washington, DC: U.S. Department of the Interior, Bureau

of Mines, RI 9135.

Signer SP [1990]. Field verification of load transfer mechanics of fully

grouted roof bolts. Washington, DC: U.S. Department of the Interior, Bureau

of Mines, RI 9301.

Signer SP [2000]. Load behavior on grouted bolts in sedimentary rock.

In: New Technology for Coal Mine Roof Support. Pittsburgh, PA: U.S. Department of Health and Human Services, Public Health Service, Centers for

Disease Control and Prevention, National Institute for Occupational Safety and

Health.

Signer SP, Mark C, Franklin G [1993]. Comparison of active versus

passive bolts in a bedded mine roof. In: Peng SS, ed. Proceedings of the 12th

International Conference on Ground Control in Mining. Morgantown, WV:

West Virginia University, pp. 16-23.

Stankus JC [1991]. Tensioned point anchor resin system vs. non-tensioned

fully grouted rebarBtwo case studies. In: Peng SS, ed. Proceedings of the 10th

International Conference on Ground Control in Mining. Morgantown, WV:

West Virginia University, pp. 65-75.
87

Figure A-1.—Map of study site at mine A.

APPENDIX.—INDIVIDUAL FIELD STUDIES

MINE A

Mine A was located in Cambria County in central

Pennsylvania [Mucho et al. 1995]. This multiple-seam

longwall mine closed in 1995 largely because of a long history

of ground control problems related to horizontal stress.

The goal of the study was to investigate the effect of a

reduced annulus on the bolt-loading history. The annulus was

reduced from 7.3 mm (0.29 in) to 3 mm (0.11 in) by installing

a No. 7 bolt 20.5 mm (0.804 in) in a 26-mm (1-1/32 in) hole

rather than the standard 35-mm (1-3/8-in) hole.

The instrumentation consisted of nine instrumented (straingauged), fully grouted, grade 75, No. 7 roof bolts installed in

the normal bolting pattern (figure A-1). As shown in the figure,

the bolts were alternately installed in reduced annulus and

normal annulus bolt holes by position in the entry. A centrally

located sonic roof extensometer and a stratascope investigation

hole were also included at the site.

Roof geology consisted of approximately 1.5 m (5 ft) of

thinly laminated shale under a coarse sandstone consisting of

0.15- to 0.6-m (6-in to 2-ft) beds with thin coal streaks

separating the beds. The CMRR value for the area was 48. The

overburden in the study area was 900 ft (275 m). 

As shown in figure A-1, a roof cutter (rock shear failure)

formed along the left (panel side) rib coincident with mining

and prior to the instrumentation or other supports being

installed. Very little roof movement was measured initially,

however. Nine days after installation there had only been a few

millimeters of movement slightly above the bolts. After several

weeks, however, the roof within the development section began

to "work" (audible noise, dripping, etc.), and workers had to be

called to the mine to set supplemental supports throughout the

section, including the instrumented area, to prevent possible

roof collapse. The extensometer showed that a movement of

slightly over 13 mm (0.5 in) had occurred above the bolts and

near the bolt top anchorage zone at the shale/sandstone interface

(figure A-2).

The bolt loads during development were a function of the

position of the bolt within the entry and relative to the roof

cutter failure. Bolts Nos. 3 and 8, in the center of the entry on

the same side as the cutter, experienced the highest peak and

average loads. Loads on these bolts exceeded the yield strength

of the steel (>200 kN (45,000 lb)) within 2 weeks of installation. Next heavily loaded were the side bolts next to the

cutter (Nos. 5 and 9). It was assumed they were slightly lower

than bolts Nos. 3 and 8 due to being installed in the failing rock

of the cutter. The reduced annulus bolts experienced loads only

slightly higher than the normal annulus bolts.

As the longwall approached, there was a total of approximately 31 mm (1.25 in) of total deformation. All locations

experienced increases in load due to the front abutment of the

longwall. Again, annulus appeared to have little effect on either

the magnitude, distribution, or timing of bolt loading.
88

Figure A-2.—Roof deformations measured at mine A.

Figure A-3.—Study site at mine B.

MINE B

Mine B was located in Greene County in southwestern

Pennsylvania [Mucho et al. 1995]. This longwall mine, now

also closed, operated in the Sewickley Coalbed. The mine also

had a long history of ground control problems associated with

horizontal stress.

The immediate roof in the instrumented area was composed

of a black shale approximately 0.3 to 0.9 m (1 to 3 ft) thick,

overlain by a dark gray shale or layered shale, sometimes with

sandy bands or grading into a silty or sandy shale. Pronounced

jointing was prevalent in the roof. Overburden in the area was

approximately 180 m (600 ft). Stress mapping determined that

the maximum horizontal stress was oriented approximately

east-west. CMRR values in the study area ranged from the high

30s to low 40s.

To evaluate the effect of mining orientation relative to the

horizontal stress field, extensometers were placed in an eastwest entry (the "good" direction) and a northwest-angled

crosscut (figure A-3). Due to severely broken ground, extensometers could not be installed in the north-south crosscuts

(the "bad" direction) as planned.

The roof in the area of the entry extensometer showed

immediate roof flaking and appeared as though it was

developing a cutter; this area was rebolted in some places along

the entry length. However, as can be seen from figure A-4, the

strains were less than 1% and confined to the lower portions of

the roof (less than the bolted interval of 2.1 m (7 ft)). Total

recorded roof deformation was only approximately 6 mm

(0.25 in), and the roof evidently was quite stable despite its

appearance.
89

Figure A-4.—Roof deformations at mine B.

The crosscut was far less stable, as evidenced by the strains

in the roof within the bolted horizon. Strains were >1% within

weeks of development at several locations. By the time the

longwall passed, the strains at 2.3 m (7.5 ft), which is above the

bolt anchorage zone, had increased to almost 4%. Total roof

deformation (sag) at that time was approximately 25 mm (1 in).

Despite the large strains, the roof was still standing after the

longwall had passed.
90

MINE C

Mine C was located in Harlan County in eastern Kentucky

[Mucho et al. 1995]. This was a drift, room-and-pillar

operation mining the Kellioka Coalbed under Black Mountain.

Overburden in the study area was 120 m (400 ft). The

immediate roof near the study area was a laminated, shaley

sandstone, often with coal streaks ("stack rock"). Calculated

CMRRs ranged between 54 and 44. 

Multipoint roof extensometers were installed in the No. 2

entry, which was supported by 1.8-m (6-ft) tensioned rebar

bolts, and in the No. 4 entry, which was supported with 1.8-m

(6-ft) fully grouted resin bolts. An adjacent crosscut was also

instrumented. The entries were oriented in the "bad" direction

approximately perpendicular to the maximum horizontal stress.

A roof cutter developed coincident with mining in the No. 4

entry (the resin bolt area), and a similar cutter developed soon

after mining in the No. 2 entry. The extensometers in the resin

bolted No. 4 entry detected roof movements within 3 hr of

installation. During the next 3 weeks, major roof movements (as

much as 7% strain) were recorded within the bolted horizon

(figure A-5). Since the cutter developed later in the tensioned

rebar area, the roof movements occurred later following mining.

The magnitudes of the movements in the tensioned rebar area

were also less than the nontensioned fully grouted bolt area,

ranging between 1% to 2% strain in the lowest 1.5 m (5 ft) of the

roof. Like the fully grouted nontensioned area, no movement was

observed above the bolted horizon during the study period.

MINE D

Mine D, located in Greene County in southwestern

Pennsylvania, was a longwall mine in the Pittsburgh Coalbed

[Mark et al. 1998; Mucho et al. 1995]. Overburden generally

ranged from 180 to 300 m (600 to 1,000 ft) at the mine. The

immediate roof is typical of Pittsburgh Coalbed geology,

alternating relatively weak shales and coals. The CMRR for the

study area was 35. The longwall panels at mine D were

oriented such that the headgate where the study was conducted

was subjected to a horizontal stress abutment.

Bolts from two manufacturers, designated "X" and "Y",

were compared in the study. Both were 2.4-m (8-ft) long,

18-mm (0.75-in) diameter, grade 75, two-piece, resin-assisted

mechanical-anchor roof bolts. The most obvious difference

between the two was that the Y bolts used 0.6 m (2-ft) of resin,

while the X version used only 0.3 m (1 ft) of resin with a

compression ring. The instrumentation plan used for the two

test sections is shown in figure A-6.

All bolts increased load shortly after installation during the

development stage; intersection bolts increased the most and

center bolts achieved the highest loads (they also were the

highest initial set loads).

The maximum load achieved by the short-anchor X bolts

averaged 84.5 kN (19,000 lb), but in many cases the load was

dropping as the longwall approached. Several Y bolts achieved

the yield limit of the steel of 150 kN (33,000 lb), and most

continued to increase their load up until the final reading as the

longwall face passed (figure A-7). Since the maximum load

achieved by the short-anchor bolts was well below the strength

of the steel, it appears that the anchors must have been slipping.

The most likely explanation is that 0.3 m (1 ft) of resin was

insufficient to maintain anchorage in this particular roof rock.

Roof strains measured during the approach of the longwall

are shown in figure A-8. At the X bolt stations, roof strains

>2% were measured at four locations within the bolted horizon.

At one intersection location, a roof strain of 6 was measured

above the bolts. The X bolts apparently began to lose control

of the ground as the horizontal stress concentration developed.
91

Figure A-5.—Roof strains measured at mine C. A, Resin bolt site; B, torque-tension bolt site.
92

Figure A-6.—Instrumentation plan at mine D.

 Figure A-7.—Roof bolt loads measured at mine D. Bolts are paired by location (i.e., the first pair shows the

intersection right-hand X bolt compared to the intersection right-hand Y bolt.
93

 Figure A-8.—Roof strains measured at mine D. The right-hand data show strains measured within the bolted

horizon; the left-hand data show strains measured above the tops of the bolts.

MINE E

Mine E was a sister operation to mine D and was similar in

most respects [Mark et al. 1998]. One significant difference

was that the horizontal stresses are relieved in the headgate by

the longwall's stress shadow. Four bolting systems were

compared in consecutive intersections at mine E:

$ Fully grouted resin bolts, 1.5 m (5 ft) long, 1.4-m (4.5-ft)

row spacing;

$ Fully grouted resin bolts, 1.5 m (5 ft long), 1-m (3-ft) row

spacing;

$ Resin-assisted mechanical-anchor bolts, 1.5 m (5 ft) long,

1.4-m (4.5-ft) row spacing; and

$ Resin-assisted mechanical-anchor bolts, 2.4 m (8 ft) long,

1.4-m (4.5-ft) row spacing.

The fourth bolting system was essentially identical to that

employed at mine E.

Very little change in roof deformation and almost no change

in bolt load was observed at any of the four sites as the longwall

approached. The maximum increase in roof strain averaged a

mere 0.2%, and all of this occurred below the bolt horizon.

Final loads on the tensioned bolts ranged between 75 and 135

kN (8 and 15 tons), considerably less than their 150-kN

(16.5-ton) yield strength. As the longwall approached, some

bolts even decreased load slightly (figure A-9). It appears that

relief of the horizontal stress may actually have enhanced roof

stability!
94

Figure A-9.—Bolt loads at mine E during the approach of the longwall.

MINE F

Mine F is a longwall mine located in Alabama's Black

Warrior Basin at a depth of approximately 670 m (2,200 ft)

[Signer et al. 1993]. Normally, the immediate roof is the

Middleman, a fossiliferous shale that grades into thinly

interbedded shale and coals. When the Middleman is the

immediate roof, mining is said to be "single-seam." The

Middleman is overlain by the 30-cm (1-ft) thick Mary Lee

Seam. The main roof above the Mary Lee consists of 30 to 61

cm (1 to 2 ft) of competent siltstone overlain by massive

sandstone. Horizontal stress caused roof guttering next to the

future longwall panel in most of the test areas. A stress

concentration was also carried with the tailgate corner during

longwall mining.

The standard primary support in the tailgate was 19-mm

(0.75-in) diameter, 1.8-m (6-ft) long, grade 40, nontensioned

fully grouted bolts. The study compared these with 22-mm

(7/8-in) diameter, 1.8-m (6-ft) long, grade 60, resin-assisted

mechanically anchored bolts. Just prior to the headgate pass, an

additional row of fully grouted resin bolts was installed

between each row of primary supports.

Four study sites were chosen, two in areas supported by fully

grouted bolts and two in areas supported by tensioned resinassisted mechanical-anchor bolts. Two sites were located in

intersections and two at midpillar. At each site, four instrumented roof bolts and three 3-point roof extensometers were

installed (figure A-10).

The data shortly after development show that there was high

localized loading in the fully grouted bolts and that several bolt

locations had reached the yield point of the steel. Generally, the

maximum load was measured by those gauges near the interface

between the Mary Lee Seam and the main roof. At the

midpillar site, greater bolt loads tended to develop on the bolts

nearest the panel rib where the cutter formed. Bolt loads in the

intersections were higher than those at the midpillar. The
95

Figure A-10.—Study site at mine F.

degree of loading on the tensioned bolts was similar to that on

the fully grouted bolts.

Between development and the headgate pass the fully

grouted bolts loaded up more rapidly than the point-anchor

bolts. Several sections of the resin bolts passed into the strainhardening phase of the plasticity curve (figure A-11). In the

final days before the tailgate passed, loads increased

significantly on nearly every bolt. Fifty percent of the straingauge stations on the resin bolts showed loads in excess of the

yield point of the steel. The load increase on the tensioned bolts

during the tailgate pass was very similar to that on the fully

grouted bolts. However, the capacity of the point-anchor bolts

was higher, and the loads remained below the yield point of the

steel.

The roof deformation measurements showed that the

greatest movements occurred along the panel rib, where the

horizontal stress guttering was present. Greater roof movements also developed at the fully grouted bolt sites during the

tailgate pass, and roof conditions were clearly more hazardous

in the fully grouted bolt areas.
96

 Figure A-11.—Bolt loads measured at mine F. The left-hand bolts are nearest the logwall panel rib. A, After the headgate pass; B, at

the tailgate corner.
97

Figure A-12.—Study site at mine G.

Figure A-13.–Maximum roof strains measured above the bolted horizon at mine G.

MINE G

Mine G employed continuous miner technology, including

mobile roof support and continuous haulage, to retreat mine the

Lower Kittanning Coalbed in central Pennsylvania [Campoli et

al. 1996]. The depth of cover varied from 45 to 120 m (150 to

400 ft). The immediate roof was a sandy shale with a CMRR

of about 50. Extensive stress mapping found local damage

related to small geologic features (slickensides and fossiliferous

bedding planes), but no significant correlation between

direction of drivage and roof damage.

Borehole extensometers were installed in the roof of seven

intersections, as shown in figure A-12. Two of the intersections

(P2 and P4) were supported by 1.52-m (5-ft) resin-assisted

point-anchor tension bolts. Three others (R2-A, R2-B, and R4)

were supported by 1.83-m (6-ft) fully grouted resin bolts. The

final two (R-P2 and R-P4) were supported by fully grouted

bolts supplemented by two additional resin-assisted pointanchor bolts between each row. Hydraulic load cells were

installed on four of the tensioned bolts at the P2 intersection.

The roof was monitored during both the developmental and

retreat mining. Roof strains approaching 1% above the bolts

were measured on development in one of the point-anchor sites

(P-2 in figure A-13), but R-P4 was not far behind. These two

intersections continued to see the greatest deformations as the

pillar line approached, probably because of nearby "cutterlike"

roof damage. Some appreciable roof deformations in the "skin"

near the roof line were observed at a number of sites as the

pillaring operations approached.

The bolt loading increased systematically from the No. 1

bolt to the No. 4 bolt. The No. 4 bolt, which was farthest from

the coal pillar and near a "cutterlike" feature, saw a load of 100

kN (23,000 lb), which is near its yield point (figure A-14). The

bolt loadings did not change significantly over time even when

the pillars were recovered.
98

 Figure A-14.—Changes in roof bolt loads measured at mine G. Bolt No. 4 is nearest a cutter; bolt No. 1 is

nearest a solid rib.
99

SKIN FAILURE OF ROOF AND RIB AND SUPPORT TECHNIQUES

 IN UNDERGROUND COAL MINES

By Eric R. Bauer1 and Dennis R. Dolinar1

ABSTRACT

Skin failures of roof and rib in underground coal mines continue to be a significant safety hazard for mine

workers. Skin failures do not usually involve failure of the support systems, but result from rock or coal

spalling from between the support elements. For instance, in 1997 more than 800 miners were injured by roof

and rib falls, of which 98% were the result of skin failures [Bauer et al. 1999]. Also, nearly 80% of the roof

and rib failure injuries occurred at or near the working faces in development sections. The face area is a zone

where the potential for skin failure accidents and injuries and for roof and rib failures is high because of mining

activity, ground readjustment due to changing stress conditions, and the higher exposure of mine workers. In

addition, failures occur where the roof and rib are unsupported. This paper reviews the roof and rib accident

statistics resulting from skin failure, and highlights the incidences by type, numbers and percentage, in-mine

location, supported and unsupported roof, and worker activity at the time of injury. Also discussed are the

causes of roof and rib skin failures, current and improved support methods and materials for skin surface

control, and machine design modifications for improved roof bolter operator protection. It also reviews the

historical literature on skin failures and control methods.

1

Mining engineer, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
100

INTRODUCTION

Falls of roof and rib traditionally have been one of the

leading causes of mine worker injuries and fatalities in

underground coal mines. From 1993 to 1998, nearly 35% of all

reported underground incidents resulted from falls of roof and

rib. These falls of roof and rib resulted in more than

4,600 injuries, or 12% of the total reported underground

injuries. Also, skin failures, which are the failure of small

blocks or slabs of roof and rib, have been recognized as a

problem in the coal mining industry for many years. Detailed

analyses showed that in 1997 alone, approximately 98% of the

roof and rib injuries were from skin failures. This suggests that

as many as 4,500 injuries may have resulted from skin failures

of the roof and rib during this 5-year period.

Reference to skin failures is found in the literature as far

back as the late 1920s. Most of the early references discussed

the effect of moisture and humidity on roof failures [Paul 1928;

Hartman and Greenwald 1941]. Other authors addressed ways

to condition mine air, such as water sprays and tempering

entries, to prevent roof deterioration [Fletcher and Cassidy

1931; Herbert 1940]. Considerable work was presented on the

effectiveness of various sealants to coat mine strata, including

coal tar [Brown 1941], Ebonol [Robbins 1937], asphalt-based

paints [Shacikaski 1951], sulfur-based coating materials [Dale

and Ludwig 1972], cement and cement mixtures [Artler 1974],

shotcrete [Cecil 1968], and polymeric sealants [Franklin et al.

1977]. More recently, researchers have investigated the

mechanisms of shale roof rock deterioration due to atmospheric

moisture, which seems to be a result of stresses from moistureinduced weakening and swelling strain, rather than slaking

[Cummings et al. 1983; Pappas and Vallejo 1997]. Finally,

although much attention has been given to the effects of

moisture and humidity on the mine roof and the resultant roof

slaking, moisture-induced skin failure is probably not the most

prevalent cause of roof skin injuries. This moisture-induced

slaking is primarily a nuisance from the standpoint of cleanup

and perception. Skin failure of the roof due to geology and

stress, in combination with mining, creates a more substantial

hazard to the miners at the face and not the long-term

deterioration of the roof due to moisture. Supporting evidence

is that nearly 80% of all roof skin injuries occur inby the feeder

breaker in development sections. To date, the problem of skin

failure at or near the working face has not been adequately

addressed. This type of skin failure will have to be addressed

by surface control systems other than sealants and by the use of

alternative methods, such as removal of a lower roof member

during mining.

Although the above literature dealt mostly with roof skin

failure, rib skin failure has also received attention by the coal

mining industry. The theory and practices regarding rib failure,

especially in thick coal seams, were addressed by Smith [1989],

who suggested that fracturing begins at a stress level equivalent

to one-third to two-thirds of the ultimate strength of the

material. Peng [1986], Dolinar and Tadolini [1991], and

Dolinar [1993] discussed general coal rib stabilization and the

effectiveness of wood dowels, resin bolts, and straps to provide

pillar reinforcement. Martin et al. [1988] provided information

that demonstrated the superior performance of yieldable rib

bolts to stabilize ribs when twin-seam mining at Jim Walters

Resources. Wykoff [1950] and Horino et al. [1971] investigated the use of wire rope to wrap pillars. Their research indicated that wire rope can significantly affect the compressive

strength and stability of pillars. In addition, many of the references on mine sealants mentioned the use of these for coating

and sealing coal ribs.

Many advances have been made in dealing with roof and rib

failures. Unfortunately, the problems have not been eliminated.

Continued research by government, academia, labor, and the

mining industry is needed to address roof and rib skin failures

and minimize the associated injuries to underground mine

workers. Research at the National Institute for Occupational

Safety and Health (NIOSH) is continuing this effort by

investigating the causes of skin failure and evaluating control

techniques.

DESCRIPTION OF SKIN FAILURE

For the purposes of this paper and analyses, skin failure does

not involve the failure of the primary support, but the spalling

of rock from between roof bolts and from around the automated

temporary roof support (ATRS) system and canopies of roof

bolting machines. Rib skin failure includes the spalling of coal

from unsupported ribs. Skin failure involves smaller pieces of

rock or coal, rather than massive roof failures (above anchorage) or coal pillar failures (bumps and bursts). Skin

failures can occur in both supported and unsupported mine

strata. Figure 1 shows skin failure of unsupported mine roof;

figure 2 is an example of skin failure of supported (bolted) mine

roof where the failure occurs between the supports. In general,

skin failure of the roof inby permanent support must be

controlled by the ATRS or canopies of the roof bolting

machine. The skin failure under permanent support can usually

be handled by removal or by surface control systems. Rib skin

failures are shown in figure 3 (unsupported rib) and figure 4

(supported rib).

The mechanisms responsible for skin failures vary

considerably. The most common factors are competence of the
101

Figure 1.—Skin failure of unsupported roof.

Figure 2.—Skin failure in supported roof.

Figure 3.—Skin failure of unsupported rib.

Figure 4.—Skin failure of supported rib.

strata and presence of geologic discontinuities. In many mines,

the roof is composed of draw rock (soft slate, shale, or rock),

coal, bony material, and other highly stratified, thinly laminated

strata. These strata are susceptible to failing in thin layers

because of bedding plane weaknesses. Some of the causes of

bedding plane failures that result in skin failures include sag of

the strata from gravity, overburden pressure as depth increases,

horizontal stress, and moisture or temperature sensitivity.

Mining-induced stresses and damage are also important in the

development of skin failure in both the roof and rib. If geologic

discontinuities are present, the likelihood of skin failure

increases because the discontinuities weaken and compromise

the structural integrity of the rock. It seems logical that the

potential for the roof to experience skin failure can be estimated

using the Coal Mine Roof Rating (CMRR). The CMRR

estimates the structural competence of coal mine roof and

considers bedding most important. It includes the factors that
102

 Figure 5.—Seam height versus injury rate for rib skin

failures, 1997 (after Bauer et al. [1999]).

 Figure 6.—Example of differential movement along a parting in

a coal rib.

weaken bedded coal-measure rocks, such as discontinuities,

moisture, and rock strength [Molinda and Mark 1994]. The

lower the CMRR, the less competent the roof and the more

susceptible it is to skin failure.

Coal ribs can experience skin failure for many of the same

reasons. Primarily, rib skin failure is associated with effects of

depth or, at times, with the early stages of the failure of

insufficiently sized pillars. A general observation about rib skin

failures is that rib spalling tends to increase as mining height

increases. A plot of rib skin failure injuries and seam height for

1997 indicates that the injury rate increased as the seam height

increased, up to 8 ft thick (figure 5). For seam heights >8 ft,

a decreasing trend occurs, probably because more rib support

is used in the thicker seams. Rib spalling may also increase

with depth and is affected by mining-induced stresses. Rib skin

failure is also frequently associated with rock partings within

the coal pillar or with draw rock located at the roof-rib

interface. Rock partings or bands within the pillar create planes

of weakness where differential movement (figure 6) and failure

can occur, leading to spalling (skin failure). When weak draw

rock that is subject to failure during coal extraction is present

and is mined with the coal, the draw rock exposed in the coal

pillars creates a zone of potential rib skin failure. This inherent

weakness makes the draw rock susceptible to spalling from the

rib as the coal pillars experience load.

SKIN FAILURE INCIDENT ANALYSIS

Two separate incident analyses were conducted. One

addressed roof and rib fall fatalities during 1996-98; the other

addressed all reported roof and rib fall incidents during 1995-

97. These analyses were designed to identify the fatalities and

injuries resulting from skin failures, both roof and rib, and

massive failures, then draw some statistically based

conclusions.

ROOF AND RIB FALL FATALITIES

The underground coal mine fatalities caused by falls of roof

and rib for the period 1996-98 were separated by skin and

massive failures. Skin failure of the roof and rib has been

previously defined. A massive roof fall involves the failure of

the primary support system and usually has an areal extent in at

least one dimension approaching the width of the opening. For

the fatalities, the average thickness of the massive failures was

8.55 ft. For rib failures, nearly all were classified as skin

failures, except those listed in Mine Safety and Health

Administration (MSHA) fatality reports as an outburst or bump.

Table 1 summarizes the classification by failure types. Essentially, 50% of the fatal injuries that occurred under

supported roof were caused by skin failure of the roof or rib.

Rib failures resulted in over twice as many fatalities as roof skin

failures and were caused by the lack of rib support, which

allows large slabs to spall from the ribs. Only three fatalities

occurred from roof skin failure; however, these occurred under

the supposedly safe conditions of supported roof. During this

3-year period, 11 fatalities occurred under unsupported roof.

This is a human behavior issue rather than a ground control

problem; thus, these fatalities are not included in this analysis.
103

Table 1.—Roof and rib fatalities by failure type,

1996-98

Year Rib skin

fatalities

Supported

roof skin

Massive

failures

1996 ..... 3 1 1

1997 ..... 3 0 6

1998 ..... 1 2 3

 Total . . . . 7 3 10

REPORTED ROOF AND RIB FALL INJURIES

To delineate the extent of worker injuries resulting from skin

failures, the MSHA accident database was examined for the

period 1995-98. All injuries occurring in underground coal

mines that resulted from roof and rib failures were extracted

and analyzed. This included degree-of-injury classes from 1 to

6, which were injuries ranging from no lost time or restricted

activity to those that resulted in a fatality. They did not include

reportable roof falls that occurred when no workers were

present. In addition, the accident injury illness types extracted

were fall of face, rib (or side), and fall of roof. Some of the

roof and rib fall injuries are classified under machinery

incidents. These misclassified incidents were sorted out by

using the source-of-injury code with a criterion of caving of

rock, coal, ore, and waste. Table 2 summarizes the roof and rib

injuries for 1995-98. The table reveals that most of the injuries

resulted from roof skin failures (82%), followed by rib skin

failures (16%), and massive failures (2%).

Table 2.—Number of injuries from

roof and rib failures, 1995-98

Failure type No. of

injuries

Percent of

injuries

Roof skin . . . . . 2,716 82

Rib skin . . . . . . 524 16

Massive . . . . . . 58 2

 Total . . . . . . 3,298 100

Next, another analysis determined the mining situations in

which roof and rib skin injuries occurred (for 1997 data only).

Table 3 indicates that 84% of the skin failure injuries occurred

during development or retreat mining, with the remaining 16%

divided among longwall and other. An attempt was made to

determine the location of skin failure injuries with respect to the

state of roof support. The best estimate is that 383 of the 669

roof skin injuries (57%) occurred under permanent support. It

is possible that many of the roof skin failure injuries occurring

where the roof was permanently supported could have been

prevented through modified support designs. Another 233

(35%) roof skin injuries occurred under temporarily supported

or unsupported roof. Increasing the skin coverage of the ATRS

or coverage area of the drill station canopies could help reduce

the roof skin failure injuries occurring under temporarily

supported roof. For the remaining 53 roof skin injuries, the

state of support was uncertain, but was provided by either the

ATRS or permanent support. Approximately 85 of the 128

(66%) rib skin injuries occurred where the roof was

permanently supported. The rib skin failure injuries occurring

under permanently supported roof may be minimized by

securing the ribs, if necessary, or through scaling and increased

awareness of rib conditions. Another 19 (15%) rib skin injuries

occurred under temporarily supported or unsupported roof. The

remaining 24 (19%) occurred where the state of the support was

unknown.

Table 3.—Roof and rib skin failure injuries classified by

mining situation, 1997

Mining situation

Roof skin failures Rib skin failures

Injuries % Injuries %

Development1 . . . . . . . 560 84 108 84

Longwall2 . . . . . . . . . . . 38 6 11 9

Other3 . . . . . . . . . . . . . 71 10 9 7

 Total . . . . . . . . . . . . 669 100 128 100

1

Includes advance and retreat mining.

2

Includes injuries in the headgate and tailgate during panel mining.

3

Includes injuries outby face and of unknown origin.

Table 4 shows the distribution of skin injuries by location

and support type. Temporary support is provided by the ATRS

and canopy of the roof bolter, while permanent support is

provided by the primary support system. About 78% of all roof

Table 4.—Location of roof and rib skin failures, 1997

Type and location Injuries Face1 Working

section2

Face area

total3

Face

area, %

Other/

unknown4 Other, %

Roof:

 Permanent support . . . . 383 150 111 261 68 122 32

 Temporary support . . . . 233 215 4 219 94 17 4

Rib:

 Permanent support . . . . 85 38 25 63 74 22 26

 Temporary support . . . . 19 16 1 17 90 2 10 1

Injuries occurring at the active face or inby the last open crosscut. 2

All other injuries occurring inby the feeder on working sections.

3

Total injuries occurring in the face and working section.

4

Injuries occurring inby the working section, or location unknown.
104

 Figure 7.—Mine worker activities during roof skin injuries,

1995-98.

 Figure 8.—Mine worker activities during rib skin injuries, 1995-

98.

 Figure 9.—Mine worker moving mining machine

power cables.

skin injuries occurred inby the feeder, while 58% of the injuries

occurred at the active face. This is a strong indicator that roof

slaking due to moisture was not the primary concern in causing

these types of injuries. Again, with the coal ribs, nearly 77% of

the injuries were inby the feeder.

Finally, the mine worker activities during roof and rib skin

injuries were extracted from the MSHA database for 1995-98.

The most common activities of workers injured by roof skin

failures were drilling or bolting of the roof (39%), operating the

continuous mining machine (11%), and general inside labor

(9%) (figure 7). These three activities accounted for 59% of the

injuries. No other worker activity was involved in more than

7% of the injuries. For injuries resulting from rib skin failures,

the most common worker activities were operating the

continuous mining machine (18%), drilling or bolting the roof

(16%), general inside labor (12%), walking (9%), and

maintenance and repair (8%). The total of these accounted for

63% of the injuries (figure 8). All other activities were involved in 5% or less of the rib skin injuries. Surprisingly,

scaling of the roof or rib, which deals directly with skin failure

and is thought to be a dangerous activity, comprised only 1% of

the roof and rib skin failure injuries. This low level of scaling

injuries compared with the high number of skin failure injuries

 may indicate that not enough scaling is done. In addition,

cable handling was involved in 3% of the total roof and rib skin

failures (figure 9).
105

SKIN CONTROL METHODS

This review of skin control methods both examines what has

been done in the past and describes current control techniques.

Our investigation reveals that many of the same methods used

in the past are used today.

EARLY SKIN CONTROL EFFORTS

Past control methods were directed primarily at preventing

skin failures resulting from changes in temperature and

humidity. These included various coating materials designed

simply to seal the surface without providing additional strength

or reinforcement and attempts to condition the mine air before

it was introduced into the mine workings. The air conditioning

involved regulating the temperature and humidity to near

ambient mine conditions to prevent failures due to expansion

and contraction and to prevent moisture variations. In the face

area, past attempts at controlling roof and rib skin failure using

artificial means reflected the support materials available.

Mechanical bolts in combination with wood headers and

planks, oversized plates, wire mesh, old hoist rope, wood

dowels, and other simple support methods were commonly

used.

CURRENT SKIN CONTROL METHODS

Current control methods have built on the successes of past

techniques, using the more sophisticated support materials now

available. In addition, more thought has been given to

matching the type of control to the failure mode. For instance,

because the mining industry has an improved understanding of

the mechanism of strata failure, cement coatings using steel or

glass fibers are available not only to seal the strata, but also to

add strength to resist failure. For control of roof skin failure,

wood planks, steel straps and channel, and various meshes such

as welded wire, chain link, or synthetic grid material are being

used.

Rib support methods have changed as well, primarily in the

use, type, and location of bolts. The emphasis is to match the

deformability of the rib supports to that of the rib. Yieldable

bolts, such as those used at Jim Walters Resources [Martin et al.

1988], can stabilize the coal seam and ribs effectively by

controlling displacements to reduce stress buildup.

A recent information request from MSHA District 3 revealed

the following examples of roof skin control methods: (1) one

mine uses screens in one intake, one return, and the track entry,

and uses a lot of gunite; (2) another mine uses 8-gauge steel,

5- by 16-ft panels of "welded wire" installed on cycle;

(3) one longwall mine is required to use screening or gunite

where it has trouble holding up head coal in its gate roads; and

(4) one mine that has a history of falls due to deteriorating top

has miles of gunited track entry. Information obtained from

MSHA District 4 revealed additional skin control methods.

These included using oversized bearing plates on pattern bolts,

installing 2-ft-long "bacon skins" (straps) with 3-ft-long

mechanical anchor bolts in between the pattern bolts or

covering the roof with synthetic grid material when roof skin

failure is a problem. For rib skin failures at the face, some

mines install 4- to 6-ft-long planks with 18- or 36-in bolts.

When sporadic rib failures occur outby the face area, mines

mainly use timbers set close to the ribs to minimize the dangers

to mine workers traveling nearby.

The following is taken from a roof control plan from a mine

in Pennsylvania, which describes typical rib skin control

methods: "Loose ribs are to be blocked, bolted, or taken down.

Steel straps, planks, or header blocks with 4- to 6-ft-long bolts

may be used. Bolts are not to exceed 8-ft intervals. In lieu of

the above, such ribs may be supported by posts or cribs

installed tightly near the rib."

ADVANCES IN SKIN CONTROL

Improved skin control and elimination of skin failure

injuries, especially those resulting from roof skin failures, are

contingent on providing increased surface control. To this end,

safer, faster, and more efficient installation of mesh is the

surface control method receiving the most attention. For

instance, the walk-through bolter allows sheets of mesh to be

installed with minimal worker exposure to the unsupported

roof. The mesh can be placed on top of the protective canopy,

then slid forward into place without the workers ever leaving

the supported roof area. The method of installing synthetic

mesh material is also being improved. An automatic grid

dispenser has been developed that mounts on the inby side of

the ATRS and dispenses the mesh up and over the ATRS

(figure 10). As the mesh leaves the dispenser, the folded edges

fan out from 9 to 15 ft in width to provide almost complete ribto-rib coverage. In addition, the mesh has been strengthened to

>13,000 psf to provide a material with similar strength and

protection as those of conventional mesh materials. Figure 11

shows the use of synthetic mesh to support roof and rib.
106

Figure 10.—Automatic grid dispenser. (Photo courtesy of Tensar Earth Technologies.)

 Figure 11.—Synthetic mesh supporting roof and rib. (Photo courtesy of Tensar Earth

Technologies.)
107

 Figure 12.—Walk-through chassis roof bolting machine. (Photo

courtesy of J. H. Fletcher and Co.)

 Figure 13.—Joy Multibolter. (Photo courtesy of Joy Mining

Machinery.)

EQUIPMENT SAFETY IMPROVEMENTS

Visits to several roof bolting machine (RBM) manufacturers

revealed that although they did not use the term "skin failure"

to describe these types of roof and rib failures, they are aware

of the problem and have been modifying the RBMs

accordingly. Most of the safety modifications involved either

removing the worker from the hazardous area or increasing the

surface coverage of protective canopies, ATRS, etc., to prevent

falling roof and rib from striking the RBM operator.

To remove the operator from the hazardous area, roof bolters

with walk-through chassis, with or without automated drill

functions, have been developed (figure 12). The major advantage of the walk-through chassis is to reduce mine worker

exposure to rib hazards. Another manufacturer's RBM, currently available for higher coal seams only, uses four automated

booms for drilling the bolt holes. The Multibolter is also a

walk-through design (figure 13). It allows the operators to

remain under a canopy equipped with side slide extensions that

provide substantial work area coverage from roof hazards. This

machine also uses side shields or chain curtain on the walkway

platform to prevent rib failure injuries.

To provide additional protection to the operator during the

bolting process, several machine modifications have been

introduced. Many of the ATRS are equipped with hydraulic or

manually extendable beams or roof contact pads to provide

more coverage between the ATRS and the rib. At least one

RBM manufacturer provides rock deflectors, called rocker

pads, on the inby side of the ATRS that deflect rocks toward the

face rather than allowing them to roll back onto the operator's

legs and feet (figure 14). This was developed in response to

injuries that occurred from dislodged rocks falling back onto the

operator when the ATRS is lowered. The deflector forces the

loose rocks to slide toward the face, falling flat against the mine

floor, rather than landing on edge and falling over onto the

operator's feet and legs. Rock deflector plates are also provided

on the ATRS boom that can help deflect falling rocks away

from the RBM operators. Another safety improvement is a

sliding extension of the drilling canopy to provide additional

surface coverage (figure 15).

Because the operator is subject to falling rocks any time that

he or she is drilling or inserting bolts, one manufacturer

developed a hydraulic resin inserter that keeps the operator

from having to reach out from under the drilling canopy.

Another improvement is to use reduced thrust, rotation, and

feed when starting to drill a bolt hole. Accident statistics have

shown that many injuries occur from falling pieces of roof rock

when bolt holes are started. Some mines have even adopted the

use of metatarsal gloves to protect the hands of RBM operators.

Ultimately, all RBM safety improvements are driven by the

desire to provide the safest work environment for the roof

bolting machine operators. Unfortunately, acceptance of these

design changes can hinge on how they affect the bolting

process. Changes that maintain the status quo or reduce bolting

cycle times are more readily adopted by the mining industry

than those that increase the time to perform any one function in

the bolting process. This is because in most room-and-pillar

operations the ability to mine the coal has outpaced the ability

to support the roof. Thus, the speed and efficiency of the roof

bolting operation is the critical production function.
108

 Figure 14.—Inby rocker pads to deflect falling rocks. (Photo

courtesy of J. H. Fletcher and Co.)

 Figure 15.—Pullout canopy extension. (Photo courtesy of 

J. H. Fletcher and Co.)

SUMMARY AND CONCLUSIONS

Skin failure of roof and ribs injures many workers in

underground coal mines. Statistics from 1997 indicate that 98%

of the injuries from roof and rib falls are due to skin failures,

resulting in more than 800 injuries, or approximately 12% of all

underground coal mine injuries.

Skin failure is defined as the failure of small pieces of rock

and coal from between the primary supports, rather than

massive roof falls or coal pillar failures. Coal ribs may not be

supported, but when the rib spalls, it is still considered skin

failure.

An analysis of roof and rib skin failure injuries revealed that

far more injuries resulted from roof skin failures, but that the rib

skin failures caused more severe injuries. The analysis also

revealed that roof and rib skin failures were three times more

likely to occur on the working section than outby in other mine

areas because of greater worker activity in the face area and because the face is an active stress zone. From a worker activity

standpoint, the roof bolters have by far the most injuries from

roof skin failure. By contrast, the risk of injury from coal rib

falls seems to be approximately the same for all face workers.

The methods for support of roof and ribs to prevent skin

failure are simply extensions of standard roof support methods.

As dictated by the extent of skin failure problems, the on-cycle

supporting methods are modified to provide additional surface

coverage. Common skin control methods include oversized

plates, header boards, wood planks, steel straps, mesh, and in

rare instances, spray coatings. These control methods can

control skin failures. Unfortunately, they are implemented reactively to control problems that are occurring, rather than

proactively to prevent future skin failure occurrences. The

success of these controls can be enhanced by matching the

characteristics of the support to the expected strata reactions to

mining and modes of failure. However, the key to preventing

injuries will be the amount of surface coverage developed by

the surface control systems.

Equipment safety enhancements, especially to the roof

bolting machine, have been directed at removing the worker

from the dangerous areas and/or increasing the area of

protective canopies. The modifications can provide additional

measures of safety to the roof bolting machine operators,

thereby reducing the potential for injuries from falling roof and

rib. Unfortunately, it is difficult to get some of the equipment

modifications adopted by the mining industry. Only those

changes that either maintain the status quo or that speed up the

bolting cycle are readily accepted, whereas other safety

modifications are more difficult to implement.

NIOSH research is continuing to address the causes and

control of skin failure in underground coal mines. Emphasis

will be placed on determining the geologic and stress conditions

associated with roof and rib skin failure and the best surface

control practices being used by the coal industry to minimize

the hazard of skin failure injuries.
109

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to stop shale degradation in coal mines. Pittsburgh, PA: U.S. Department of

the Interior, Bureau of Mines, TPR 103.

Hartman I, Greenwald HP [1941]. Effects of changes in moisture and

temperature on mine roof. First report on strata overlying the Pittsburgh

Coalbed. Pittsburgh, PA: U.S. Department of the Interior, Bureau of Mines,

RI 3588.

Herbert CA [1940]. Cooling mine air during summer months to prevent

roof falls. Pittsburgh, PA: U.S. Department of the Interior, Bureau of Mines,

IC 7098.

Horino FG, Duvall WI, Brady BT [1971]. The use of rock bolts or wire

rope to increase the strength of fractured model pillars. Pittsburgh, PA: U.S.

Department of the Interior, Bureau of Mines, RI 7568.

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Resources, mining division. In: Peng SS, ed. Proceedings of the 7th

International Conference on Ground Control in Mining. Morgantown, WV:

West Virginia University, pp. 66-75.

Molinda GM, Mark C [1994]. Coal mine roof rating (CMRR): a practical

rock mass classification for coal mines. Pittsburgh, PA: U.S. Department of

the Interior, Bureau of Mines, IC 9387.

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nondurable shales associated with coal mine waste embankments. In: Hudson

JA, ed. Proceedings of the 36th U.S. Rock Mechanics Symposium. New

York, NY: Columbia University, paper No. 241.

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seasons and rate of production). Pittsburgh, PA: U.S. Department of the

Interior, Bureau of Mines, TP 410.

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NY: John Wiley & Sons, pp. 417-420.

Robbins VC [1937]. New roof treatment proves out at southwestern mines.

Coal Age 42(5):219.

Shacikaski A [1951]. Coating for roof protection. Coal Age 56(12):88-90.

Smith WC [1989]. Evaluation of progressive rib failure in thick coal

seams. SME preprint 89-174. Littleton, CO: Society for Mining, Metallurgy,

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187(8):898-902. New York, NY: American Institute of Mining, Metallurgical,

and Petroleum Engineers.
111

DESIGN OF ROOF BOLT SYSTEMS

By Christopher Mark, Ph.D.1

ABSTRACT

Roof bolt system design means the selection of the type, length, capacity, and pattern of bolts for a

particular application. Despite research efforts dating back 50 years, no design methodology has found wide

acceptance. This paper begins by identifying four mechanisms that roof bolts use to reinforce the ground. It

argues that the reinforcement mechanism is determined by the roof geology and stress level, not by the type

of bolt. Next, the attributes of roof bolts are discussed in the light of recent research, including anchorage

mechanism, installed tension, length, capacity, timing of installation, and installation quality. Several

significant areas of controversy are identified. Design methods from around the world are discussed, including

those based on empirical research, numerical modeling, and roof monitoring. Finally, some simple guidelines

for preliminary design of roof bolt systems are proposed based on statistical analysis of roof support

performance at 37 U.S. mines.

1

Supervisory physical scientist, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
112

INTRODUCTION

Roof bolts work with the ground to create a stable rock

structure. They are the first line of defense to protect mineworkers from the hazards of ground falls. Because roof bolts

use the inherent strength of the rock mass, they have many

advantages when compared with earlier standing support

systems. Roof bolts were first introduced in the United States

shortly after World War II and quickly became the dominant

mode of roof support. Resin-grouted systems represented

another improvement over mechanical bolts and have been increasingly favored since the 1970s. As other countries have

adopted high-production retreat longwall methods, roof bolting

has spread internationally. Roof bolts largely supplanted steel

sets first in Australia in the 1970s and 1980s and then in the

United Kingdom and Canada during the 1990s. Currently,

Germany and other European coal-producing countries are

adopting them [Martens and Rattmann 1999].

Because of their central importance, roof bolts have received

more research attention than any other ground control topic, with

the possible exception of coal pillars. Numerous roof bolt design

methods have been proposed, but a recent survey paper

concluded that none "has gained any acceptance by the coal

mining industry" [Fuller 1999]. It seems that the complexities of

the bolt-ground interaction continue to defy complete solution.

Nevertheless, some important knowledge can be gleaned

from the mass of available literature. This paper presents the

state-of-the-art as it applies to reinforcement mechanisms, roof

bolt attributes, and design methodologies. Some simple guidelines for roof bolt selection are then proposed.

REINFORCEMENT MECHANISMS OF ROOF BOLTS

The principal objective of roof bolting is to help the rock

mass support itself. Some researchers have ascribed different

support mechanisms to different types of roof bolts. For example, mechanical bolts were originally thought to work in

suspension, whereas resin bolts primarily built beams [Gerdeen

et al. 1979]. Others have described the beam-building mechanism of tensioned bolts, and the frictional support of fully

grouted bolts [Peng 1998].

It seems, however, that the reinforcement mode is actually

dictated to the bolts by the ground, rather than the reverse. The

degree of reinforcement required and the principal reinforcement mechanism depends on the geology and the stress regime.

Four levels of support, each using a different support mechanism, can be identified:2

$ Simple Skin Control: Strong, massive roof subjected to

low stress levels can be essentially "self-supporting," meaning

that a major roof collapse is unlikely to occur. However,

cracks, joints, crossbeds, or slickensides can create occasional

hazardous loose rock at the skin of the opening (figure 1A).

Pattern bolting is therefore required to prevent local loose rock

from falling, but the bolts may be relatively short and light.

Skin control is also an important secondary function of roof

bolts with the other three support mechanisms.

$ Suspension: In many mines, a thin layer of weak,

immediate roof can be suspended from an overlying thick,

strong unit that is largely "self-supporting" (figure 1B). 

 2

It is interesting to note that Thomas, in 1954, listed the same first three

mechanisms of roof bolt support, although his definitions varied somewhat

from the ones given here.

Experience has shown that roof bolts are extremely efficient in

the suspension mode [Conway 1948; Damberger et al. 1980;

Mark et al. 1994b], although suspension becomes more difficult

if the weak layer is more than 1 m (3 ft) thick. The Coal Mine

Roof Rating (CMRR) somewhat quantifies this effect through

the Strong Bed Adjustment [Molinda and Mark 1994].

$ Beam Building: Where no "self-supporting" bed is within

reach, the bolts must tie the roof together to create a "beam"

(figure 1C). The bolts reinforce the rock by maintaining

friction on bedding planes, keying together blocks of fractured

rock, and controlling the dilation of failed roof layers [Peng

1998; Gale et al. 1992]. In general, it is much more difficult for

roof bolts to build a beam than it is to suspend weak rock from

one.

$ Supplemental Support Required: Where the roof is

extremely weak or the stress extremely high, roof bolts alone

may not be sufficient to prevent roof failure from progressing

beyond a reasonable anchorage horizon (figure 1D). In these

cases, cable bolts, cable trusses, or standing support may be

necessary to carry the dead-weight load of the broken roof, and

the roof bolts act primarily to prevent unraveling of the

immediate roof [Scott 1992].

In practice, these mechanisms are not always clearly defined.

In particular, the transition between suspension and beam

building depends heavily on the level of stress. A roof bed that

is "self-supporting" when subjected to low stress may require

reinforcement when the stresses increase. Wider spans also

reduce the self-supporting ability of the roof [Mark and Barczak

2000]. Figure 2 summarizes the concepts presented here.
113

 Figure 1.—Roof support mechanisms. A, simple skin support; B, suspension; C, beam building; D, supplemental support in failing

ground.

 Figure 2.—Roof support mechanisms determined by stress level and roof

quality.
114

 Figure 3.—Stiffness of fully grouted and resin-assisted pointanchor bolts compared (using data from Karabin and Hoch [1980]).

 Figure 4.—Tension in a fully grouted bolt caused by dilation of

a failed roof bed.

 Figure 5.—Tension and bending in a fully grouted bolt caused

by slip on a bedding plane.

CHARACTERISTICS OF ROOF BOLTS

Roof bolts are defined by a number of characteristics, including anchorage mechanism, installed tension, length, etc.

The relative importance of these individual attributes have

sometimes been the subject of much controversy.

Anchorage MechanismBPoint-Anchor Bolts: Two basic

types of anchorage are available: point-anchor and fully

grouted. Mechanical shells are the older type of point anchors,

but these have now largely disappeared from U.S. mines

[Dolinar and Bhatt 2000]. Today, resin-assisted mechanical

anchor bolts are often used to support difficult conditions. 

Point-anchor bolts carry high loads at the anchor and at the

collar, but do not contact the rock over most of their length.

Since they must be installed with tension, their initial stiffness

is "infinite" until the rock load exceeds the initial tension.

However, because their further response to any rock movement

is distributed along their entire length, the stiffness of pointanchor bolts is lower than that of fully grouted bolts [Karabin

and Hoch 1980] (figure 3).

Pullout tests are the standard technique for determining the

anchorage capacity of point-anchor bolts. The anchorage is

considered adequate if it exceeds the breaking strength of the

bolt. If the anchorage is found to be inadequate, it may be

improved in a number ways [Mazzoni et al. 1996]. Because

point-anchor bolts that lose their installed tension are almost

entirely ineffective, Federal regulations at 30 CFR 75.204

require that they be tested. Anchor creep was the biggest

problem with mechanical bolts, but this is seldom a problem

with resin-assisted point-anchor bolts. Roof deterioration at the

plate is another concern, and wooden headers should be avoided

because they can creep under load and shrink as they dry.

Anchorage MechanismBFully Grouted Bolts: Fully grouted

bolts are loaded by movement of the rock. The movement may

be vertical sag, shear along a bedding plane, or dilation of a

roof layer buckled by horizontal stress (figures 4-5). The

move-ments cause tensile forces in the bolt, often combined

with bending stresses [Signer 2000; Fabjanczyk and Tarrant

1992]. Figure 6 shows typical load distributions in a fully

grouted bolt.

The stiffness of a fully grouted bolt is determined by the

load-transfer mechanisms between the rock, the grout, and the
115

 Figure 6.—Typical load distributions measured in a fully

grouted bolt at three time during its service life (after Gale [1991]).

Figure 7.—A short encapsulation pull test.

bolt. Signer [1990] provides an excellent discussion of load

transfer mechanisms. Good load transfer exists when very high

loads develop in the bolt in response to small ground

movements, and these loads are rapidly dissipated away from

the zone of roof movement. Poor load transfer can result in

[Fabjanczyk and Tarrant 1992]:

$ Large plate loads;

$ Larger roof movements before maximum bolt response;

and

$ Lower ultimate bolt capacity, particularly if roof

movements occur near the top of the bolt.

One way of expressing the effectiveness of load transfer is

the "bond strength." Bond strength is actually a misnomer

because there is no adhesion between the resin and the rock,

just mechanical interlock [Karabin and Debevic 1976]. In this

paper, the term "anchorage factor" will be substituted for "bond

strength." The anchorage factor is obtained from short

encapsulation pull tests (figure 7), in which the grouted length

is short enough that the anchorage fails before the bolt yields

[Karabin and Debevic 1976; Health and Safety Executive

1996]. The anchorage factor, in kilonewtons per millimeter or

tons per inch, is determined by dividing the applied pulling load

by the anchorage length. Typically, no more than 300 mm (12

in) of the bolt is grouted in a short encapsulation test, and tests

may be conducted at a variety of depths to evaluate the load

transfer characteristics in different roof beds. Standard pullout

tests should not be employed with full-length resin bolts

because the pulling forces seldom extend more than 450 mm

(18 in) up the resin column [Serbousek and Signer 1987].

Table 1 gives typical anchorage factors and anchorage

obtained from the literature. Short encapsulation tests are

apparently rather rare in the United States; the only available

published data were obtained from Peng [1998]. Although the
116

 Figure 8.—Effect of hole annulus on grouted bolt performance.

Results were obtained from short encapsulation pull tests on

19-mm (0.75-in) diameter rods (after Karabin and Debevic [1976]).

Table 1.—Anchorage factors for fully grouted resin bolts

Rock type Country Anchorage factor,

N/mm (tons/in)

Length for 90 kN

(10 tons) of

anchorage,

mm (in)

Coal, shale . . . . . . . . . . . . . . Australia . . . . . 300- 900 (0.7-2.1) 100-300 (4-12) 

Hard sandstone, limestone . . Australia . . . . . 1,000-2,500 (2.3-5.8) 35- 90 (1.4-3.6)

Minimum allowable . . . . . . . . U.K. . . . . . . . . . 400 (1.1) 225 (8.9)

Soft rock . . . . . . . . . . . . . . . . U.S.A. . . . . . . . 180 (0.5) 510 (20) 

Strong rock . . . . . . . . . . . . . . U.S.A. . . . . . . . 720 (2) 125 (5 ) 

Australian data [Yearby 1991] and the U.K. data [Bigby 1997]

probably apply to slightly larger bolts, there seems to be a clear

difference. The implication is that in weak rock in the United

States, the top 500 mm (20 in) or more of a fully grouted bolt

may required to develop an anchorage force equal to the

breaking strength of the rod. In such conditions, the "effective

capacity" of the upper portion of the bolt may be considerably

less than its nominal capacity.

A number of factors can affect the load transfer characteristics and anchorage factor, including—

Rock Strength: Weaker rock requires a longer grouted

length to achieve the same anchorage capacity as strong rock

[Franklin and Woodfield 1971; Karabin and Debevic 1976].

One study of the former U.S. Bureau of Mines [Cincilla 1986]

found that coal and shale roofs required an average of 800 mm

(31 in) of grouted length to achieve full anchorage, while

sandstone required 460 mm (18 in) and limestone needed just

300 mm (12 in). In very weak rock, anchorage factors can be

so low that 1.6-m (6-ft) bolts have been pulled from the rock at

14 tons even though they were fully grouted for their entire

length [Rico et al. 1997].

Hole annulus: Numerous tests over the years have found

that optimum difference between the diameter of the bolt and

the diameter of the hole is no greater than 6 mm (0.25 in),

giving an annulus of about 3 mm (0.125 in) [Fairhurst and

Singh 1974; Karabin and Debevic 1976; Ulrich et al. 1989].

For example, a 3-mm (0.125-in) annulus is obtained by a

19-mm (0.75-in) bolt in a 25-mm (1-in) hole. Results from

short encapsulation pull tests on 19-mm (0.75-in) bolts are

shown in figure 8.

Larger holes can result in poor resin mixing, a greater

likelihood of "finger-gloving," and reduced load transfer

capability. One Australian study found that the load transfer

improved more than 50% when the annulus was reduced from

4.5 to 2.5 mm (0.35 to 0.1 in) [Fabjanczyk and Tarrant 1992].

Smaller holes, on the other hand, can cause insertion problems

and magnify the effects of resin losses to roof cracks or to

overdrilled holes [Campoli et al. 1999]. However, one recent

U.S. study found that annuli ranging from 2.5-6.5 mm (0.1-0.25

in) all provided acceptable results in strong rock [Tadolini

1998]. Also, if failure is occurring at the resin-rock interface in

very weak rock, increasing the hole diameter is one way to

decrease the shear stress on the interface [Rico et al. 1997].

Hole and bolt profile: Because resin grout acts to transfer

load by mechanical interlock, not by adhesion, rifled holes and

rougher bolt profiles result in better load transfer [Karabin and

Debevic 1976; Haas 1981; Aziz et al. 1999]. Reportedly, wet

drilled or water-flushed holes can also improve load transfer

[Siddall and Gale 1992]. One study found that the pullout load

of standard rebar was seven times that of a smooth rod

[Fabjanczyk and Tarrant 1992].

Resin characteristics: Tests in the United Kingdom in the

late 1980s demonstrated that the compressive strength of resin

was important to the performance of grouted roof bolts [British

Coal Technical Department 1992], and current U.K. regulations

require resin strength to exceed 80 MPa (11,000 psi).

A strength test was recently added to the American Society for

Testing and Materials (ASTM) standards for resin. However,

an extensive series of laboratory "push tests" found little

correlation between shear stress and resin strengths in the 20-60

MPa (3,000-6,000 psi) range [Fabjanczyk and Tarrant 1992].
117

C ' π

4

GD 2 (1)

Figure 9.—The Euler buckling beam concept (after Frith [1988]).

In summary, good load transfer is essential for optimizing

the performance of resin bolts, particularly in weak rocks. U.S.

mines have been criticized for using "vacuum drilling, large

diameter holes, and low strength resin" [Hurt 1992]. Although

field measurements indicate that U.S. resin bolts usually

respond quickly to roof movements, which indicates good load

transfer properties [Signer and Jones 1990; Signer et al. 1993;

Maleki et al. 1994; Signer and Lewis 1998], low anchorage

factors may reduce the effective capacity of the upper portion

of bolts installed in some weak rock conditions. It may be

possible to improve bolt performance by adjusting load transfer

properties such as hole size or rifling. More widespread use of

short encapsulation pull tests (figure 8) could be very helpful in

identifying when and where low anchorage factors may be a

problem.

INSTALLED TENSION

One of the most controversial topics in roof bolting is the

importance of installed tension. Numerous papers have been

written pro and con in Australia and the United States. The

issue can be further confused because there are actually three

possible systems: fully grouted nontensioned, fully grouted

tensioned, and point-anchor tensioned.

In the United States, Peng [1998] argues that resin-assisted

point-anchor tensioned bolts can be used to clamp thinly

laminated roof beds into a thick beam that is more resistant to

bending. Stankus and Peng [1996] add that by "increasing

frictional resistance along bedding planes, roof sag and

deflection is minimized, and lateral movement due to horizontal

stress is unlikely to occur." Tensioned bolts are also said to be

more efficient, because "a stronger beam can be built with the

same bolt by utilizing a larger installed load."

Frith and Thomas [1998] advocate pretensioning fully

grouted bolts using two-stage resins and special hardware.

They argue that active preloads modify roof behavior by

dramatically reducing bed separation and delaminations in the

immediate 0.5-0.8 m (2-3 ft) of roof. A key reason that tension

works, they say, can be understood if the roof is seen as an

Euler buckling beam. Small vertically applied loads therefore

have a mechanical advantage that allows them to resist high

horizontal forces (figure 9). Fuller [1999] concludes that "the

generally positive results of field trials indicates that pretensioning when combined with full bonding of bolts provides

the maximum strata reinforcement."

Gray and Finlow-Bates [1998] put the case that nontensioned, fully grouted bolts with good load transfer

characteristics may be just as effective. They argue that a

preload of 100 kN (12 tons) results in a confining stress of only

70 kPa (10 psi) on the roof, which is minimal compared with in

situ horizontal stresses which are at least 100 times greater.

Also, the loads dissipate rapidly into the rock. Others have

observed that in field measurements, resin bolts have quickly

achieved loads that are even greater than those on nearby pointanchor bolts [Mark et al. 2000]. McHugh and Signer [1999]

found that in laboratory tests, the confining loads applied by

pretensioned, fully grouted bolts did little to strengthen rock

joints.

Unfortunately, direct comparisons of the three systems are

relatively rare. Anecdotal evidence is often cited, sometimes

from situations where bolt length and capacity were changed as

well as tension [Stankus 1991]. There is general consensus that

large preloads are not necessary for resin bolts to function

effectively in the suspension mode [Peng 1998; Frith and

Thomas 1998; Maleki 1992], but broader conclusions apparently must wait for more research.

It should be pointed out that fully grouted bolts are not

entirely tension-free. In the United States, there is typically

about 11 kN (1 ton) of plate load when the bolts are installed

[Signer 1990]. Plate loads can increase by a factor of 10 or

more in highly deforming ground [Tadolini and Ulrich 1986].

The thrust bolting technique can apply upwards of 44 kN

(4 tons) of initial plate load [Tadolini and Dolinar 1991], which

is similar to what is measured on the typical Australian

"nontensioned" roof bolt [Frith and Thomas 1998].

BOLT CAPACITY

The yield capacity (C) of a roof bolt is normally determined

by the bolt diameter (D) and the grade of the steel (G):

For rebar, the diameter is usually given as a number, where

#5 rebar is 5/8 in (16 mm) in diameter, #6 is 0.75 in (21 mm),

and so on. The grade of the steel is normally given in thousands

of psi, where a grade 40 steel is 40,000 psi (280 MPa), etc. The

grade and the diameter, and some other information including

the bolt length, are stamped on the head of the bolt, using the

symbols shown in table 2.
118

BL ' S

2

100 & RMR

100 [Unal 1984] (4)

BL ' S

3 [Bieniawski 1987] (3)

Table 2.—Markings on the heads of roof bolts

(ASTM F432-95, “Standard Specification for Roof and Rock Bolts and Accessories”)

Headed

bolts

Nominal product

size, in

Mfg.

symbol1 Diameter2 Grade3 Length, in

GR 40 . . . . . . . . . 3/4 and over Yes Yes None Yes

GR 55 . . . . . . . . . 5/8 and over Yes Yes t Yes

GR 60 . . . . . . . . . 5/8 and over Yes Yes a Yes

GR 75 . . . . . . . . . 5/8 and over Yes Yes X Yes

GR 100 . . . . . . . . 5/8 and over Yes Yes ~ Yes

 1

Enter alpha-numeric symbol. 2Enter numerical value of bolt diameter measured in eighths of an inch; numerical value of

deformed bars placed in circle.

 3

Grades above 100 are produced in 20-ksi increments; they are marked 2 for 120 ksi, etc.

The ultimate capacity of a bolt is often considerably greater

than the yield. Table 3 shows yield and ultimate capacities for

several common bolts. In general, lower grade steels are more

ductile than high-strength steels, meaning that there is a

relatively greater difference between the yield and the ultimate

strength. Signer [1990] points out that while a typical rebar will

yield after 0.8 mm (0.030) in of deformation, an additional

50 mm (2 in) is required to break it.

Table 3.—Load-carrying capacities of mine roof bolts

Roof bolt material Minimum yield,

MPa (psi)

Minimum ultimate

tensile, MPa (psi)

5/8 Grade 55 . . . . . . . 86 (12,400) 132 (19,200)

5/8 Grade 75 ........ 117 (17,000) 156 (22,600)

3/4 Grade 75 ........ 173 (25,100) 230 (33,400)

#6 Rebar Grade 40 . . . 121 (17,600) 212 (30,800)

#6 Rebar Grade 60 . . . 182 (26,400) 273 (39,600)

#7 Rebar Grade 40 . . . 166 (24,000) 290 (42,000)

#7 Rebar Grade 60 . . . 248 (36,000) 372 (54,000)

#5 Rebar Grade 60 . . . 127 (18,600) 190 (27,900)

Several factors may cause the actual bolt capacity to be

somewhat less than the capacity of the rod. The most obvious

is if the anchorage is inadequate. Although all bolts must be

tested to ensure that they meet ASTM specifications, coupled

bolts are sometimes prone to fail at the coupler. Poor

installation can also cause a stress concentration at the bolt

head. In thin seam mines, bolts are sometimes notched so that

they can be bent more easily. The cross-section area of the steel

left in the notch then determines the bolt capacity. In general,

notches rolled into the bar reduce strength less than machined

notches.

Many authors argue in favor of greater capacity to improve

the effectiveness of roof bolts [Gale 1991; Stankus and Peng

1996]. One obvious advantage is that stronger bolts can carry

more broken rock. Higher capacity bolts are also capable of

producing more confinement and shear strength in the rock, and

they may be pretensioned to higher levels. Larger diameter

bolts are also stiffer.

The increased capacity may not be utilized in all circumstances, however. Field studies show that bolts are not loaded

equally, and the roof may fail on one side of the entry before the

bolts on the other see significant loads (figure 10). More

importantly, if the roof is failing above the bolts, it may fall

without ever loading them. On the other hand, if broken bolts

are observed in roof falls, increased bolt capacity is clearly

indicated.

BOLT LENGTH

The optimal roof bolt length depends on the support

mechanism. Where bolts are merely acting as skin control, they

may be as short as 750 mm (30 in). In the suspension mode,

bolts should obtain at least 300 mm (1 ft) of anchorage in the

solid strata. Federal regulations at 30 CFR 75.204 require that

when point-anchor bolts are used, test holes must be drilled at

least another 300 mm (1 ft) above the normal anchorage.

In some mines, the thickness of the weak, immediate roof

layer can vary by as much as 1 m (3 ft) over very short

distances. In these mines, roof bolt crews select the proper

length bolt based on their observations while drilling. They

sense where they contact the strong bed from the sound and

penetration rate of the drill. Computerized feedback control

technologies are now being developed which may aid drill

operators in identifying strong anchorage horizons [Thomas and

Wilson 1999].

The proper bolt length is more difficult to determine in the

beam-building mode. Some empirical rules of thumb that have

been suggested include:

BL ' S2/3 [Lang and Bischoff 1982] (2)

where BL ' bolt length;

 S ' span; and

 RMR ' rock mass rating [Bieniawski 1987].
119

Figure 10.—Nonuniform bolt loading measured in an entry developed at an angle to the maximum horizontal stress [Gale 1991].

The Unal equation is the most appealing of the three because

it considers the rock quality in addition to the span (note that the

CMRR may be substituted for the RMR in equation 4). The

Unal equation was not intended for intersection spans, however,

nor does it consider stress level. None of the three equations

have been validated for use in coal mines.

It seems that increasing bolt length can be a very effective

measure for reducing roof falls. The study reported by Molinda

et al. [2000] found that out of 13 mines where 2 different lengths

of bolts were used in similar roof conditions, the fall rate was

lower for the longer bolts 84% of the time. The same study found

little support for the theory that shorter bolts installed at higher

than normal tensions can reduce roof fall rates [Stankus and Peng

1996]. It should be noted, however, that the effective capacity of

the upper portion of a fully grouted bolt can be significantly

reduced if the load transfer is poor, whereas a resin-assisted pointanchor bolt should function along its entire length (as long as the

length of the resin column is adequate).

As equations 2 through 4 suggest, wider spans require longer

bolts for beam building. In coal mines, the widest spans are

generally found in intersections. However, most mines use the

same length bolt both in intersections and entries. This may

help explain why intersections are as much as 10 times more

likely to collapse (on a foot-per-foot basis) than entries

[Molinda et al. 1998]. Many mines that are experiencing high

rates of roof falls might be able to improve conditions by using

longer bolts just in intersections.

ROOF BOLT PATTERN

The density of roof bolt support varies little in the United

States. With the advent of dual-head roof bolting machines,

four bolts per row has become the near-universal standard. Bolt

spacing is limited by law to a maximum of 1.5 m (5 ft), but is

seldom <1.2 m (4 ft). With entries varying in width from about

4.5-6 m (15-20 ft), bolt densities range from approximately one

bolt per 2.4 m2

 (25 ft2) to one bolt per 1.4 m2 (15 ft2).

Such patterns are appropriate for the vast majority of U.S.

applications, which are for simple skin control, suspension,

and beam building at relatively low stress. By international

standards, however, they are quite light for beam building in

high-stress conditions. In the United Kingdom, the minimum

bolt density allowed by statute is one bolt/m2 (11 ft2

), and

many Australian mines use similar bolt densities. In these

countries, higher bolt densities are considered necessary to

maximize the strength of failed rock around the roadways

[Gale et al. 1992]. The lighter patterns used in the United

States may help explain why some mines have such difficulty

controlling the weakest roof in highly stressed ground.

Unfortunately, higher bolt densities are probably not

economically feasible in the United States, primarily because

of their impact on drivage rates.

One partial alternative that might be helpful in some cases is

to put extra bolts in where the bolts are most heavily loaded.

The field study reported by Maleki et al. [1994] found that

increasing the bolt density reduced the average bolt load, while

the total load remained approximately the same. Other researchers have found that when one side of the entry suffers

greater stress damage, bolts on that side receive significantly

more load [Mark and Barczak 2000; Siddall and Gale 1992].

Additional bolts on the stress-damage side can help maintain

overall stability.
120

 Figure 11.—Effect of the depth of cover on the stability of

unsupported roof.

TIMING OF BOLT INSTALLATION

As soon as a cut is mined, the roof begins to move. Some

relaxation is necessary to relieve the in situ stress, but excessive

movement can reduce the strength of the rock mass by reducing

the confinement on bedding planes and other discontinuities.

The longer a roof remains unbolted, the more likely that some

damage will occur.

The degree of potential damage depends on the stress level,

the span, and the roof quality. Whereas strong roof may not

suffer at all, weak roof under high stress may collapse before

the miner completes the cut. The study of extended cuts

reported by Mark [1999] found that when the CMRR exceeded

55, extended cuts were nearly always stable. In these conditions, very little damage apparently occurs before the bolts are

installed. When the CMRR was between 55 and 40, most

mines had mixed experiences with extended cuts, indicating that

the roof tends to degrade with time and should be bolted soon

after mining. Deeper mines also had more trouble than

shallower ones, indicating that elevated stresses also require

quick support (figure 11).

The study also found that mines with a CMRR < 38 could

rarely employ extended cuts. Place-change mining, which

requires that the roof stand unsupported until the bolting

machine arrives, may not be economic under such conditions.

The difficulties in place-change mining highly stressed, weak

roof explains the prevalence of miner-bolters in the Pittsburgh

Seam. Miner-bolters are single-pass machines that mine a

narrower entry and allow the roof to be bolted minutes after it

is exposed. Pittsburgh Seam mines have found that roof fall

rates are reduced substantially with this mining method. Most

mines in Australia and the United Kingdom use similar systems.

SKIN SUPPORT

Skin support is an essential function of roof bolt systems,

serving the dual purposes of—

• Protecting miners from small rocks that could fall between

the bolts; and

• Preventing the roof from unraveling and ultimately

negating the purpose of the bolts.

Skin support is achieved through a combination of plates,

headers, mats, straps, mesh and sealants. Skin support is the

subject of a current research study under the National Institute

for Occupational Safety and Health (NIOSH). Some preliminary results are reported by Bauer and Dolinar [2000].

INSTALLATION QUALITY

Poorly installed support is, at best, ineffective and, at worst,

provides a false sense of security. Unfortunately, it is difficult

to check the installation of most modern roof supports.

Whereas timber supports can be checked visually and

mechanical bolts can be checked with a torque test, resin

anchors have thus far defied attempts to develop an effective

testing technique.

The troubleshooting guide prepared by Mazzoni et al. [1996]

provides the most complete information available on roof

support quality. The guide attributes problems with roof bolts

to three main sources:

• Geology;

• Poor installation quality; and

• Defective support hardware.

With fully grouted bolts, potential installation problems

include—

• Defective grout due to improper storage, improper

temperature at the time of installation, or manufacturing error;

• Defective hole due to crookedness, cracks, improper

length, or improper diameter;

• Poorly mixed grout due to improper insertion, rotation,

thrust, torque, spin time, or hold time; and

• Defective bolt. Tensioned grouted systems can suffer from

all of the problems listed above, as well as defective couplers,

shear mechanisms, threads, washers, and anchors.

The miners who operate roof bolt machines are the key to

maintaining high-quality support installations. Certainly, there

is no substitute for job training and experience. In addition,

knowledge about strata reinforcement principles can be very

effective in motivating roof bolt crews to ensure quality support

throughout the mine [Fuller 1999].
121

P ' U( t ( We

( R

n % 1 SF , (5)

 Figure 12.—Failure mechanisms of a fully grouted bolt [after

Serbousek and Signer 1987]. A, roof movement near head; B, roof

movement in central portion; C, roof movement in anchorage

zone.

ROOF BOLT FAILURE MECHANICS

Roof bolts can fail in one of several ways:

• The head or the plate can fail;

• The rod may break, either in tension, or a combination of

tension and bending; or

• The anchorage may fail.

In addition, roof bolts may be intact, but the support system

can fail if—

• The bolts are too short, allowing the roof to fail above

them; or

• The bolts fail to provide adequate skin control, allowing

loose rock to create a hazard or letting the roof unravel over

time.

Point-anchor bolts normally fail by anchor slip or by

exceeding the capacity of the steel. A sudden break can cause

the freed bottom end to be released at high speed [Peng 1998].

This hazard is known as the "shotgun effect."

Studies have shown that a very high percentage of resin bolts

are loaded to their yield point, sometimes very early in their

service lives [Signer 2000]. Data presented by Signer [1990]

seem to indicate that once the steel yields, it pulls away from the

grout, greatly reducing the load transfer that takes place along

that portion of the bolt. If the lower portion of the bolt yields,

it can be manifested as increased plate loads (figure 12A).

Loading in the central portion may ultimately break the rod

(figure 12B). However, anchorage failure may occur if there is

poor load transfer near the top of the bolt, whether caused by

bolt yielding or not (figure 12C). Considering the anchorage

factor data presented in table 1, if a typical U.S. roof bolt

installed in weak rock was loaded in its upper 500 mm (20 in),

it could be pulled out of the hole before the rod yielded.

Once a standard roof bolt is loaded to its ultimate capacity,

it usually has very little residual strength. Compared with many

supplemental supports (e.g., wood cribs and cable trusses), roof

bolts are normally effective over a relatively small range of

deformation. However, there is a class of yielding roof bolts

that are designed to maintain high loads through deformation

ranges of 300 mm (12 in) or more. Yielding bolts normally

employ a slip-nut at the bolt head. They are designed for very

high deformation environments, such as long-term applications

in creeping salt, or pillarless longwall extraction under extremely deep cover [Terrill and Francke 1995; VandeKraats et

al. 1996; 1998; Martens and Rattmann 1998].

APPROACHES TO THE DESIGN OF ROOF BOLT SYSTEMS

Various methods for the design of roof bolts have been

proposed through the years. None has achieved wide success.

Today, most roof bolts are still selected using a combination of

past experience, trial and error, and regulatory requirements.

Much can still be learned from a review of the different

concepts. The survey below briefly describes a number of

theories, an approximately chronological order. The bolt design

attributes that they address are also identified.

Dead-weight design (capacity/pattern): The oldest,

simplest, and probably still most widely used equation for bolt

design is dead-weight suspension [Obert and Duvall 1967]:

where P ' required bolt capacity;

U ' unit weight of the rock;

t ' thickness of suspended rock;

n ' number of bolts per row;
122

ht ' B 100 & RMR

100 (6)

Rock mass strength ' Uniaxial compressive strength

K , (7)

Figure 13.—Dead-weight loads on roof bolts.

Figure 14.—The rock load height concept (after Unal [1984]).

We ' entry width;

R ' row spacing; and

SF ' safety factor.

Figure 13 gives dead-weight loads calculated for various bolt

spacings. This method is probably suitable for suspension

bolting in low-stress environments. However, horizontal forces

can greatly increase the loads applied to roof bolts [Wright

1973; Fairhurst and Singh 1974]. Signer et al. [1993] found

that measured loads on roof bolts are often twice what would be

predicted by dead-weight design.

Rock Load Height (capacity/pattern): The rock load height

concept is a slightly more sophisticated version of the

deadweight theory. Originally proposed by Terzaghi [1946],

the theory predicts the load on the supports based on the rock

quality and the span. Unal [1984] defined the rock load height

for coal mining:

The rock load height is illustrated in figure 14. Again, the

CMRR may be substituted for the RMR in equation 6.

Panek's Chart (length/tension/pattern): An early attempt at

a comprehensive design procedure was presented by Panek

[1964]. He conducted a series of scale model tests using

limestone slabs to represent roof beds. His results were

presented in the form of a nomogram that related bed thickness

and roof span to the required bolt length, tension, and pattern.

Remarkably, Panek's nomogram continues to be republished,

although it is very doubtful that it has been used for practical

design in decades [Fuller 1999].

Other Physical Models (location): In the prenumerical

modeling era, several researchers used physical models to

explore roof bolting performance [Fairhurst and Singh 1974;

Dunham 1976; Gerdeen et al. 1979]. All of these studies

assumed that the roof was perfectly bedded, and they

consistently found that bolts located in the center of the entry

added little to roof stability. In contrast, one model study of

roof containing low-angle shears as well as bedding found that

an evenly spaced pattern performs best [Mark 1982].

Peng and Guo (pattern): Peng and Guo [1989] used a

hybrid boundary-element/finite-element model to design the

spacing for fully grouted bolts. The models incorporated weak

bedding planes, and parametric analyses were performed in

which roof stiffness, layer thickness, and horizontal stress were

varied. By applying dimensional analysis, they derived a series

of equations that give the number of bolts required to prevent

bed separation, tensile fracture, shear fracture at midspan, and

shear fracture at the entry corners. Some simple guidelines for

bolt length were also presented.

Two-Phase Ground Support (support type/timing): Scott

[1992] proposed that when longwall entries that are expected to

undergo large deformations, a two-phase ground support system

might make sense. The first phase would consist of short,

closely spaced rock anchors that would slip at their loadcarrying capacity, but continue to prevent the immediate roof

from unraveling as it deformed. The second phase would

consist of long cable anchors or standing supports capable of

carrying the weight of the fractured ground while accepting its

dilation. Scott cited the gabion analogy in support of his theory.

Scott's approach could result in a more efficient design than one

that tried to prevent all deformation, and it can be argued that

many U.S. longwalls that install heavy standing support in the

tailgate already use a version of it.

Maleki (bolt type): Maleki [1992] proposed a preliminary

criterion for bolt selection based on his analysis of 20 case

histories. The factors determining the type of bolt required are

the stress level and the rock mass strength. The laboratory rock

strength is downgraded to give the rock mass strength as

follows:
123

Figure 15.—Maleki’s [1992] roof bolt selection chart.

where K ' 1 for massive strata; K ' 2 for cohesive, medium

bedded strata; and K ' 3 for finely laminated, noncohesive

strata (figure 15).

In Maleki's approach, tensioned, fully grouted bolts are

recommended for the most difficult conditions.

Design by Measurement (pattern/length): This design

approach was developed in Australia [Gale 1991; Gale and

Fabjanczyk 1993] and was largely adopted by the U.K. Code of

Practice [Bigby 1997]. The basic concept is that as individual

roof beds become overstressed and fail, they force stresses higher

into the roof, which can in turn fail more beds (figure 16).

Reinforcement aims to mobilize the frictional strength of failed

roof beds in order to restrict the height and severity of failure in

the roof. It involves measuring the loads developed in roof bolts

during mining, together with a definition of the height and

severity of roof deformation obtained from multipoint extensometers. Based on the measurements, optimization of the

bolting design might include—

• Adjusting the bolt length so that adequate anchorage is

achieved above the highest level in the roof where failure is

occurring;

• Adjusting the bolt density and placement to maximize

reinforcement where the roof needs it most;

• Improving load transfer by reducing hole size, optimizing

bit type, or flushing the hole.

The results are considered valid for environments that are

similar to the one studied. Significant changes in the geology

or stress field requires additional monitoring.

Optimum Beaming Effect (tension/length): Stankus and

Peng [1996] proposed the Optimum Beaming Effect, which is

defined as the roof beam that has no separation within or above

the bolted range and uses the shortest bolt possible. Its basic

tenet is that high installed tensions can be substituted for bolt

length. They also argue that longer bolts elongate more in

response to load, therefore allowing more roof deformation.

The method has been implemented in a finite-element model

(see section on "Numerical Modeling" below). Unfortunately,

there does not seem to be sufficient justification for this theory.

Molinda et al. [2000] found that shorter, tensioned bolts had

higher roof fall rates than longer, nontensioned ones in three of

four cases where both bolts were used in the same mine.

Structural Engineering Model (tension): In Australia, Frith

[1998] proposed a model that divides mine roof into two

classes:

• Static roof that is essentially self-supporting and requires

minimum reinforcement; and

• Buckling roof that is thinly bedded and tends to fail layerby-layer due to horizontal stress.
124

 Figure 16.—Failure sequence in highly stressed roof (after Gale

[1991]).

Frith proposes that the behavior of the second type of roof

can be explained by the basic structural engineering concept of

the Euler buckling beam (see previous section on "Installed

Tension"). There have been a number of trials of high-tension

fully grouted bolts in Australia, and the results are reported to

be positive. Unfortunately, the field evidence that has been

presented to date has been largely anecdotal (see, for example,

Rataj et al. [1997]).

Numerical Modeling: As computers and software have

grown more powerful, numerical modeling has become the

standard design tool in many branches of engineering. Rock

mechanics, however, has lagged behind. The reason is that rock

engineers cannot specify the properties of the materials that they

use, nor can they usually define the their loading conditions

adequately.

For effective, quantitative design using numerical models,

three basic prerequisites must be met [Hayes and Altounyan

1995; Gale and Fabjanczyk 1993]:

• Model: The model must be capable of replicating the

behavior of coal measure rock, which means it must be able to

simulate the various failure modes and large deformations

which typically occur.

• Material Properties and Stress: Input rock mass properties

must reflect both pre- and postfailure mechanics of the different

roof layers encountered, and in situ stress levels must be measured in the field.

• Validation: To ensure that the model and the ground are

behaving the same way, stresses and displacements must be

measured. Important parameters include the magnitude and

location of deformations, the distribution of bolt loads, and the

behavior of interfaces at the top of the pillar and within the roof.

Numerical models used in the United States seldom meet any

of these requirements. Stankus and Guo [1997] and Guo and

Stankus [1997] describe a finite-element model that uses gap

elements every 300 mm (1 ft), but otherwise assumes the rock

to be homogeneous, elastic, and isotropic. The model looks for

zero separation within or above the bolted range, which the

study's authors cite as a weakness because bedding separations

are commonly observed underground even where the roof is

adequately supported [Stankus and Guo 1997]. The model's

results are also extremely sensitive to the frictional strength

coefficient [Guo and Stankus 1997]. The movements predicted

by the model also seem quite small. In one instance cited by

Stankus and Peng [1996], the total modeled roof deflection was

<1 mm (0.032 in), and the longest bolt resulted in just 6% more

deformation than the shortest.

Rigorous models that seem to meet all of the necessary requirements for quantitative design have been described overseas

[Bigby 1997; Gale and Tarrant 1997]. Such models implement

as many as seven rock failure modes, including bedding slip,

shear failure of intact rock, tensile failure, and buckling. However, the expenses associated with such elaborate models,

including the associated rock testing, stress measurement, and

monitoring, are probably beyond customary U.S. practice.

Moreover, the rapid changes in geology that often occur

underground raise the question of the number of models and

verification sites that might be needed.

Fortunately, numerical models can be very valuable tools even

if there is not enough information to use them for quantitative

design. As Starfield and Cundall [1988] pointed out, models can

be used as controlled experiments to investigate the qualitative

effects of different parameters. Well-designed model studies could

be very helpful in moving the science of roof bolting forward.

ROOF MONITORING

Regardless of roof bolt design, failures are always possible.

Often, an unstable area can be controlled with secondary support if the problem is detected in time. In the United States, instability is usually detected from visible and audible signals that

become apparent shortly before collapse. Instruments are far

more sensitive and can detect ground movements much earlier.

Routine monitoring of roof movements is much more common abroad. In the United Kingdom and Canada, two-point

extensometers (often known as "telltales") are required every

20 m (65 ft) in bolted roadways and in all intersections

(figure 17). The telltales have two movement indicators, one

that shows displacement within the bolted height, and the other
125

Figure 17.—A telltale (after Altounyan et al. [1997]).

that shows movement above the bolts. Telltales are visible to

everyone using the roadway, and their information can be

recorded for later analysis [Altounyan et al. 1997].

The key to the effective use of monitoring is the

determination of appropriate "action levels." For example, in

gate roads at the Phalen Mine in Nova Scotia, Canada

[McDonald and McPherson 1994]:

• Spot bolting when 25 mm (1 in) of movement is recorded

either within or above the bolts.

• Additional bolting and center props when 50 mm (2 in) of

movement is recorded.

• Cable bolts when 75 mm (3 in) of displacement is observed.

In the United Kingdom, typical action levels are 25 mm

 (1 in) within the bolted horizon and 10-25 mm (0.4-1 in) above

[Kent et al. 1999a]. A survey of action levels in Australian

mines, however, found no such uniformity. Some mines used

total movement criteria; others used rates of movement ranging 

from 1 to 10 mm (0.04 to 0.4 in) per week [Mark 1998]. In the

United States, the data are scarce, but action levels or "critical sag

rates" have usually been about 5 mm (0.2 in) per week [Mark et

al. 1994c].

In the United States, the lack of available personnel to

install, read, and interpret roof monitors has always hindered

their widespread use. However, preventing even a single roof

fall in a critical belt or travel entry could justify the expense of

a fairly extensive monitoring program. Hopefully, the time is

not far away when computerized systems will help mines to

make better use of roof monitors.

Often, roof monitoring can uncover a hidden geologic factor

that can then be used directly in design. For example, a back

analysis of monitoring data from the Selby coalfields in the

United Kingdom found that excessive roof movements occurred

where entries were unfavorably oriented relative to the horizontal stress or where the mudstone thickness exceeded

2.5 m (8 ft) [Kent et al. 1999b]. At the Plateau Mine in Utah,

Maleki et al. [1987] found that excessive sag rates correlated

with the presence of a channel sandstone within 1.5-2.2 m

(5-7 ft) of the coal. A program of test holes helped locate the

sandstone and reduced the number of sagmeters needed.

 GUIDELINES FOR ROOF BOLT DESIGN

Currently, there are no reliable methods for designing roof

bolt systems. To begin to fill the mining community's need for

better guidelines, NIOSH conducted a study of roof fall

frequencies at 37 coal mines. The study's methodology, data

collection procedures, and statistical analyses are reported by

Molinda et al. [2000].

The study found that there was considerable scatter in the

results, so that it was not possible to develop a universal design

equation. In particular, it was not possible to determine the

relative importance of individual rock bolt parameters including

tension, length, capacity, and pattern.

However, some valuable relationships were found. It was

not surprising that the geology, represented by the CMRR, was

the most important variable. However, the next most important

parameter was the depth of cover. With all else equal, deeper

mines were more likely to have high roof fall rates. Horizontal

stress could not be measured directly, but since it is known that

the intensity of horizontal stress tends to increase with depth,

the inference is that the depth of cover is a surrogate for the

stress level. When the data were separated into a shallow cover

group (<125 m (400 ft) and a deeper cover group (>125 m

(400 ft), bolt design equations were determined for each.

Following are step-by-step guidelines:

1. Evaluate the geology. The CMRR should be determined

either through underground observation or from exploratory

drill core. Zones of markedly different CMRR should be

delineated. If the thickness of individual beds varies within the

bolted horizon, this effect should be noted. Special features,

such as faults or major geologic transition zones, should be

treated separately.
126

LB ' 0.12 Is log10 (3.25H) 100 & CMRR

100 (meters) (9a)

LB ' 0.12 Is log10 (H) 100 & CMRR

100 (feet) (9b)

PRSUP ' Lb ( Nb ( C

Sb ( We (10b)

PRSUP ' 29 Lb ( Nb ( C

Sb ( We (10a)

2. Evaluate the stress level. It is unusual for stress

measurements to be available, so the design procedures use the

depth of cover as a rough estimator. However, horizontal stress

can sometimes be intensified by stream valleys or by driving in

an unfavorable orientation. Roof support may need to be

increased in these areas.

3. Evaluate mining-induced stress. Vertical, and sometimes

horizontal, stresses may also be intensified by retreat mining or

multiple seam interactions. These areas are likely to require

supplemental support.

4. Determine the intersection span. An equation was

derived from the data which suggests that the appropriate

diagonal intersection span (Is) is approximately:

(Is) ' 9.5 + (0.2 ( CMRR) (meters) (8a)

(Is) ' 31 + (0.66 ( CMRR) (feet) (8b)

If the CMRR > 65, it should be set equal to 65 in equation 8.

The intersection span can also be estimated from the entry

width using table 4 where the typical spans are based on the

field data:

Table 4.—Diagonal intersection spans (Is)

Entry width,

m (ft)

Ideal span,

m (ft)

Typical diagonal

intersection spans

Shallow cover,

m (ft)

Deep cover,

m (ft)

4.9 (16) . . . . . 7.0 (23) 8.9 (29) 9.5 (31)

5.5 (18) . . . . . 7.8 (25) 9.5 (31) 10.1 (33)

6.2 (20) . . . . . 8.7 (28) 9.8 (32) 10.5 (34)

 NOTE: The "ideal span" is determined by applying the Pythagorean

theorem (a2 % b2 ' c2

). "Typical" spans are based on actual measurements [Molinda et al. 2000].

As table 4 shows, the field data indicated that for the same

entry width, spans at deep cover (depth > 130 m (400 ft))

exceeded the shallow cover spans by an average of 0.6 m (2 ft)

due to pillar sloughing.

5. Determine the bolt length. Where the roof geology is

such that the suspension mode is appropriate, the bolt length

should be selected to give adequate anchorage in the strong

rock. For the beam building mode, a bolt length formula was

derived by modifying the Unal [1984] rock load height

equation. The intersection span was substituted for the entry

width, a depth factor was added, and then the constant was

adjusted to fit the data:

where (Is) ' diagonal intersection span (meters in equa-

 tion 9a; feet in equation 9b); and

H ' depth of cover (meters in equation 9a; feet in

 equation 9b).

These equations are illustrated in figure 18.

6. Determine bolt pattern and capacity: As has already

been stated, the data could not determine which bolt parameter

was most important. Therefore, the design variable is PRSUP,

which includes both, plus the bolt length:

where Lb ' length of the bolt (meters in equation 10a; feet

 in equation 10b);

 Nb ' number of bolts per row; 

C ' capacity (kilonewtons in equation 10a; kips in

 equation 10b);

Sb ' spacing between rows of bolts (meters in

 equation 10a; feet in equation 10b); and 

We ' entry width (meters in equation 10a; feet in 

 equation 10b).

Note that PRSUP differs from the PSUP used in past studies

[Mark et al. 1994a] in that the bolt capacity has been substituted

for the bolt diameter.

The suggested value of PRSUP for shallow cover is

determined as:

PRSUP ' 15.5 & 0.23 CMRR (11a)

and for deeper cover:

PRSUP ' 17.8 & 0.23 CMRR (11b)
127

Figure 18.—Formula for selecting the bolt length. A, depth = 1,200 ft; B, depth = 800 ft; C, depth = 300 ft.
128

 Figure 19.—Design equations for selecting bolt pattern and capacity. The field data used in the derivation of

the formulas are shown, along with the original “discriminate equations” (dotted line). A, shallow cover (depth

< 120 m (400 ft); B, deep cover (depth > 120 m (400 ft)).

Figure 19 shows these equations, together with the field data

from which they were derived. The design equations are

slightly more conservative than the discriminant equations on

which they are based.

The field data also indicated that in very weak roof, it may

be difficult to eliminate roof falls using typical U.S. roof bolt

patterns. When the CMRR was <40 at shallow cover and

<45-50 at deeper cover, high roof fall rates could be encountered, even with high roof bolt densities. Faced with these

conditions, special mining plans, such as advance-and-relieve

mining (Chase et al. [1999]), might be considered.

It should also be noted that these equations have been

derived to reduce the risk of roof falls in intersections. In some

circumstances, it may be possible to reduce the level of support

between intersections.

Finally, the minimum recommended PRSUP is approximately 3.0.

7. Select skin support: Plates, header, mats, or mesh should

be specified to ensure that loose rock between the bolts does not

pose a hazard.

8. Monitoring: The installation of telltales or other simple

extensometers should be considered for critical intersections so

that, if it becomes necessary, supplemental support can be

installed in a timely fashion.
129

CONCLUSIONS

1. Four support levels and reinforcement mechanisms are

identified for roof bolts: simple skin control, suspension, beam

building, and supplemental support required. The mechanism

required for a particular application depends on the geology and

the stress level.

2. The performance of fully grouted roof bolts can be

determined by the load transfer effectiveness, which is indicated

by the anchorage factor. Poor load transfer can reduce the

effective capacity of the upper 300-600 mm (1-2 ft) of the bolt.

Installations in weak rock are most at risk. Short encapsulation

tests can be used to determine if the Anchorage Factor is

adequate. Load transfer can be improved by optimizing the hole

annulus, rifling or cleaning the hole, or roughening the bolt

profile.

3. The importance of installed tension remains a subject of

controversy. High tension is probably not necessary for simple

skin control or suspension applications, but it may be helpful

for beam building.

4. Increasing the bolt length can be effective in reducing the

number of roof falls.

5. In weak roof, it is important that roof bolts be installed as

soon as possible after the roof is exposed.

6. Effective skin control is an essential function of all roof

support systems.

7. Proper installation is critical to the performance of roof

bolting systems. Unfortunately, it is difficult to check the

installation of fully grouted systems. Training and retraining of

roof bolt crews is therefore essential.

8. Roof bolts may fail at the head, in the rod, or at the

anchor. In addition, the system may fail if the rock breaks

above it or if the support does not provide effective skin

control.

9. Field measurements have shown that the loads on roof

bolts commonly exceed the dead-weight loads by factors of two

or more. Unfortunately, most of the other available empirical

design approaches are qualitative at best.

10. Before numerical models can be used for design, they

must—

• Be sophisticated enough to replicate complex rock mass

behavior;

• Incorporate detailed rock property and in situ stress data;

and

• Be validated by extensive field measurements.

Models used in the United States rarely meet these criteria.

11. Roof monitoring, particularly with two-point extensometers, could greatly improve our capacity to optimize the performance of roof bolt systems in the United States. However,

such instruments will have to be computerized before they are

widely accepted by the mines. A better understanding of the

appropriate "action levels" for U.S. conditions will also be

needed.

12. Guidelines are suggested for the preliminary design of

roof bolt systems, based on analysis of field data collected from

37 U.S. coal mines. Formulas are provided that may be used to

select appropriate intersection spans, bolt lengths, and bolt

capacity/patterns. The formulas require a determination of the

roof quality (using the CMRR) and the stress level (using the

depth of cover). The equations should be used with caution,

however, because the data used in their derivation were highly

scattered.

13. The data also suggest that typical U.S. bolting systems

may not always be capable of controlling roof falls in weak rock

subjected to high stress.

14. Much more progress is needed before roof bolt design

can truly be said to have advanced from an "art" to a "science."

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Aziz N, Indraratna B, Dey A, Wang Y [1999]. Laboratory study of shear

loading and bolt load transfer mechanisms under constant normal stiffness

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133

DESIGN METHODOLOGY FOR STANDING SECONDARY

ROOF SUPPORT SYSTEMS

By Thomas M. Barczak,1 Thomas P. Mucho,2 and Dennis R. Dolinar3

ABSTRACT

Maintaining ground stability in the gate roads, particularly the tailgate, has always been critical to the

success of longwall mining, both in terms of safety and productivity. Several new support technologies have

been developed in recent years to replace conventional wood and concrete cribbing for secondary roof support.

Since their performance characteristics are unique, the best practices that have been developed with

conventional wood cribbing may not be applicable for these alternative support technologies. Therefore, with

so many options to consider and the importance of achieving adequate ground control at minimal cost, the trialand-error approach to longwall gate road support is no longer prudent. This paper discusses a design

methodology for standing secondary tailgate supports. This design technique requires in-mine measurements

of tailgate support loading and convergence to establish a tailgate ground reaction behavior based on support

and strata interaction. The methodology uses the performance characteristics generated in the National

Institute of Occupational Safety and Health's (NIOSH) Mine Roof Simulator (MRS) to match the stiffness and

load characteristics of various supports to the measured ground reaction behavior. It can be used to determine

the appropriate application of alternative roof support systems or to design in-mine trials so that a fair and

equitable comparison of different support systems can be made. A case study of the methodology at a western

Pennsylvania mine site is presented in the paper, including a comparison of four alternative support

technologies to the conventional wood and concrete cribbing historically used at this particular mine.

1

Research physicist.

2

Branch chief, Disaster Prevention and Response.

3

Mining engineer.

Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
134

INTRODUCTION

The longwall tailgate support system must flawlessly control

the tailgate ground conditions. Safety considerations, especially the limited escape routes from a longwall face, demand

that the tailgate entry be a negotiable travelway. The location

of the face electrical systems, support equipment, and belt line

in the head gate entries dictate that the tailgate may be the only

option for mine workers to escape from the face in the event of

an emergency. A recent example is the longwall gob fire that

occurred at a mine in Utah in 1998 in which several miners

evacuated through the tailgate entry to safety. In addition to the

emergency travelway requirements, inadequate tailgate support

that results in poor ground control and blocked tailgates due to

roof falls can severely retard or halt production. The heavy

reliance by mines on the longwall production for survival

dictates that loss of production for protracted time periods

cannot be tolerated. Ventilation is another issue that depends

on proper tailgate support. As the panel lengths continue to

increase, excessive closure or restriction of the tailgate entry by

deformation and/or density of the standing support can be

problematic and potentially unacceptable. In gassy mines, it

also may be required that the tailgate be kept open inby the

longwall face in order to establish effective bleeder ventilation

of the tailgate area. Another important issue to consider is the

material handling aspects of tailgate supports. Therefore, the

onus is on mine engineers to design a support system that

maintains adequate control of the tailgate ground conditions at

all times and with minimal ventilation resistance and material

handling considerations.

Historically, the importance of ground control has led to very

conservative applications of tailgate support. Most mines use

conventional wood crib structures. When properly designed,

conventional wood cribs provide effective ground control in

most longwall tailgate entries and in the past have been cheap

enough that mines could afford to use a high density of cribs at

relatively little cost. However, increasing timber costs, inconsistent timber quality that has led to poor crib performance,

and inadequate supplies of timber for Western mines have reduced the advantages of conventional wood cribbing and have

encouraged many mines to consider other options for tailgate

support. All of these factors have prompted support manufacturers to develop innovative support technologies as alternatives to conventional wood cribbing.

Today, there are several alternative support technologies that

have been developed by various support manufacturers and

tested for safety and performance characteristics at the National

Institute for Occupational Safety and Health’s (NIOSH) Safety

Structures Testing Laboratory. With increasing pressures to reduce support costs while ensuring the safety of the mine

workers, mine operators more than ever before are looking for

ways to optimize longwall tailgate support. The performance

characteristics of these supports are unique, so the best practices

that have been developed for conventional wood cribbing

largely through trial and error may not be applicable for these

alternative support technologies. In addition, these new supports have limitations, which, if not properly recognized, can

lead to poor installation and inadequate tailgate ground control.

Therefore, with so many options to consider and the importance of achieving proper ground control at minimal cost,

the trial- and- error approach to longwall tailgate support can be

costly and indecisive. This paper proposes a design philosophy

for longwall tailgate secondary support whereby the ground

reaction behavior, as a function of support load density and

stiffness of the support system, is determined by measurements

of underground support loading and roof-to-floor convergence.

The goal of support design is then to optimize the use of the

support by designing to the ground reaction curve and controlling convergence to acceptable limits that will ensure

stability of the mine roof. This approach will allow mine operators to maximize the use of alternative support systems while

ensuring the safety of mine workers by avoiding risky and timeconsuming trial-and-error assessments of support technologies.

In addition, it will provide Mine Safety and Health Administration (MSHA) with a means to assess various support

systems on an equivalent basis when approving roof control

plans. A case study is included in the paper that relates the use

of the proposed design methodology to a trial of alternative

support technologies at a western Pennsylvania longwall mine

operating in the Pittsburgh coal seam. 

TERMINOLOGY

Secondary Support - Secondary support is support that is

intentionally added to assist the primary support (roof bolts) in

controlling the mine roof when it is known that additional roof

loading will occur. In longwall mining, secondary support is installed in advance of abutment loading. Secondary support is not

to be confused with supplemental support, which is support

installed in addition to primary and secondary support either for

insurance purposes or in response to unanticipated poor ground

conditions.

Ground Reaction Curve - A concept of how the ground reacts

to the presence of a newly created opening. Specifically, as the

ground deforms and sheds load to other structures, there will be

a proportional decrease in roof loading and required support

capacity to maintain equilibrium of the mine roof and floor. 

Critical Convergence - In relation to the ground reaction curve,

critical convergence is the point where failure of the ground is
135

inevitable, and the full weight of the failed rock mass above the

mine entry must be supported by the secondary roof system to

prevent a roof fall. The goal of secondary support design is to

prevent this convergence.

Support Load Density - The load-carrying capacity of an

installed support system per unit area of exposed ground

(tons/ft2

) at a particular amount of ground deformation. 

Minimal Acceptable Support Load Density - Lowest load

density of support that should be provided. A lower support

load density would allow convergence greater than critical

convergence and thereby allow failure of the roof rock that may

lead to a roof fall.

Support Density - Term typically used to refer to the number of

supports per unit area. Support density should not be confused

with support load density, which is the capacity of the support

system per unit area as a function of convergence.

Support Stiffness - A measure of how quickly a support

develops its load capacity in relation to convergence. For an

individual support, stiffness can be determined from the slope

of the load-deformation performance curve. "Softer" supports

have a flatter slope than "stiffer" supports when plotted to the

same load-displacement scale. Softer supports require more

convergence to develop an equivalent load-carrying capacity

than stiffer supports.

Support System Stiffness - The resistance to load of a group of

supports. System stiffness is the sum of the stiffnesses of individual supports. Hence, a double row of supports would have

twice the system stiffness of a single row of the same type of

supports.

KEY FACTORS TO CONSIDER IN GATE ROAD SUPPORT 

A design philosophy for standing supports must be based on

the interaction of the support with the surrounding rock mass.

The question that needs to be addressed to formulate a design

methodology is to determine to what extent the support system

is controlling the ground. To do this, it is necessary to understand both ground behavior and the characteristics of both the

individual support and the support system.

UNDERSTANDING THE FUNCTION OF THE

SUPPORT SYSTEM

Obviously, the support system is employed to prevent roof

falls. How this is accomplished is the important issue. While

secondary supports provide the last means of support in the

event there is roof failure above the bolted horizon, the primary

function of the secondary support system is to assist the primary

support system in maintaining the integrity of the immediate

roof. As the ground deforms by the creation of an opening

during mining, it gradually sheds load to the surrounding mine

structures, which, in the case of longwall mining, are the gate

road pillars and the longwall panel. Secondary support must be

placed in sufficient time and develop sufficient capacity to

bring deformation of the ground into equilibrium before a

critical deformation is reached, at which point failure of the

ground is inevitable. Otherwise, the secondary support will be

required to carry the entire dead weight of the detached rock

mass to prevent a roof fall. This embodies a fundamental

concept in rock mechanics known as the "ground reaction

curve"[Deere et al. 1970].

In longwall mining, the tailgate entry is subjected to three

phases of loading and equilibrium. Each will have a distinct

ground reaction curve. The first phase occurs on development

where the mine opening is created and the primary support

(roof bolts) is installed. Relatively little ground movement

takes place during this phase since the development loads are

small and the primary support is sufficient to provide

equilibrium. The next phase is adjacent panel mining. The

future tailgate is subjected to side abutment loading, and while

secondary support is typically installed to ensure that

equilibrium of the rock mass is obtained, the convergence is

typically minimized by the load density of the support. 

The final phase of tailgate behavior is where the active

tailgate is subjected to front abutment loading from panel

extraction. It is this phase where the secondary supports play

their most important role in preserving the stability of an entry.

A hypothetical tailgate ground reaction curve is shown in figure 1. It should be noted that the ground reaction curve will be

a function of several factors in addition to the load density of

the support system. These include geology, roof spans, vertical

and horizontal stress around the opening, and some timedependent factors such as creep. Hence, the ground reaction

curve is generally unique to a specific mine and can change

within the mine as these factors change. From the perspective

of secondary support design, it is important that the ground

reaction curve be examined under worst-case load conditions

where ground control is required. Since ground reaction is

dependent upon roof span, a different ground reaction behavior

will typically be observed in intersections as compared to the

nonintersection areas of the entry. Hence, the support design

must be altered for the intersections to accommodate this

difference in ground reaction.

It is seen from the hypothetical tailgate ground reaction

curve (figure 1) that if the goal is to prevent convergence

completely, then the full abutment load must be resisted by the
136

Figure1.–Ground reaction curve concept.

support system. For all practical purposes, this resistance cannot be obtained by installation of secondary support since the

required capacity would need to be equivalent to that of all the

coal removed by the mining process. The support capacity

required to achieve equilibrium is reduced as deformation increases, since the roof is shedding load to other mine structures

as it deforms. In other words, by allowing the roof to deform

and shed some load to the coal pillars and longwall panel, less

support capacity is required because the roof load is decreased.

Hence, the lowest required support capacity would be one that

is developed just before critical roof deformation reaches the

point where failure of the immediate roof is fast approaching.

However, designing to this lower limit of support capacity

leaves no margin of error in the event that load conditions

worsen. Also, it can be seen that if a support system is too soft

(develops load-carrying capacity too slowly), equilibrium of the

mine roof will never be achieved and failure of the roof will be

inevitable.

In summary, since most standing supports are passive supports, convergence must take place before sufficient support

loads are developed to provide equilibrium. While some supports have active loading capability, the magnitude of active

loading is not sufficient to achieve equilibrium or the active

loading cannot be maintained indefinitely. The supports must

be loaded in compression to provide the required load-resisting

forces to achieve equilibrium of the rock mass. The amount of

convergence required to produce equilibrium is then a function

of the stiffness of the support system. Equilibrium will be

achieved at less displacement for a stiffer support system because in a stiff support system, resistance to roof loading will

develop quicker (at less displacement) than in a softer support

system. Hence, as the support load resistance (load density) of

a support system increases, the convergence at which

equilibrium is attained will decrease. If too much convergence

is permitted through use of too soft a support system, failure of

the rock mass (mine roof) will be inevitable. Hence, the goal of

support design is to provide sufficient support stiffness to ensure that the required support capacity to achieve equilibrium of

the rock mass occurs before the rock mass deforms to the point

of failure. However, a prudent mine engineer would ensure that

sufficient support capacity is developed long before critical

convergence is reached. Since minimizing convergence is

achieved by increasing the support capacity (load density) (and

generally the cost of support), the goal of optimizing support

selection is not to install more support than is necessary to

provide a reasonable margin of safety to prevent roof failure. 

UNDERSTANDING THE PERFORMANCE

CHARACTERISTICS OF THE SUPPORT

The load-displacement characteristics of numerous roof support technologies have been determined at the Safety Structures

Testing Laboratory through full-scale tests in the unique Mine

Roof Simulator (MRS) load frame [Barczak 1994]. The loaddisplacement response of these various support systems are

documented in figures 2 through 7, grouped by the following

description of the support type: (1) conventional wood cribbing

(figure 2), (2) engineered wood crib supports (figure 3),

(3) conventional and engineered timber post supports (figure 4),

(4) nonyielding concrete supports (figure 5), (5) deformable

concrete supports (figure 6), and (6) yielding steel supports

(figure 7). 

As previously indicated, all secondary supports must be

loaded in compression to produce the required capacity to

achieve equilibrium of the mine roof. In other words, the roof

has to move down before the standing support develops

sufficient load-carrying capacity to achieve equilibrium of the

mine roof and floor. Since it is this very downward movement

of the roof that we are trying to control, the most important

design parameter for standing supports is the stiffness of the

support system. Stiffness is simply a measure of how quickly

a support develops its load-carrying capability in response to

convergence of the mine roof and floor. Stiffer supports develop equivalent load carrying capacity with less displacement

than softer support systems. 

While the stiffness of the support is the primary design

parameter, it is not the only parameter to consider in support

application. Another important design parameter is the loadcarrying stability of the support. More specifically, it is important to know how well the support can sustain its loadcarrying capability as a function of convergence. Stiff supports,

such as the nonyielding concrete support (figure 5), which develop load-carrying capacity quickly, but fail at little convergence, are not practical in many longwall tailgate applications. To keep such supports from failing prematurely, a

large number of supports must be installed so that roof loading

is sufficiently shared among several supports while achieving
137

Figure 2.–Load-displacement performance data for conventional wood.

Figure 3.–Load displacement performance data for engineered wood crib supports.
138

Figure 4.–Load-displacement performance data for conventional and engineered timber props.

Figure 5.–Load-displacement performance data for nonyielding concrete support.
139

Figure 6.–Load-displacement data for deformable concrete supports.

Figure 7.–Load-displacement performance data for yielding steel supports.
140

Load density (wood cribs) ' 4 cribs × 40 tons/crib

16 × 16 ft

' 0.625 tons/ft 2

(1)

Load density (concrete cribs) ' 4 cribs × 65 tons/crib

16 × 12 ft

' 1.3 tons/ft 2

(2)

Load density (2&in convergence) ' &0.20171 × 2.0 % 1.45

' 1.04 tons/ft 2

(3)

high enough load density to keep convergence below the failure

point for any one support.

The ideal support is one in which stiffness can be controlled

through the support design so that the support capacity at a

given displacement can be engineered to match ground reaction

behavior. Furthermore, the ideal support would be able to

maintain this capacity through a wide range of displacements

without shedding load or failing prematurely and therefore

would provide a margin of safety in the event that ground

conditions worsen unexpectedly.

MEASUREMENT OF THE GROUND REACTION

CURVE AT A WESTERN PENNSYLVANIA MINE

Studies were made at a longwall mine operating in the

Pittsburgh coal seam in western Pennsylvania. The mine has

historically used a staggered double row of four-point wood

cribs on 8-ft spacings (figure 8) for longwall tailgate support.

Due to uncertainties and inconsistent timber qualities, the mine

also employed concrete cribs constructed from normal mine

ventilation stopping blocks. These were also arranged in a

staggered double row, as shown in figure 9. Since the stiffness

of these two supports are significantly different (figures 2 and

5), and the supports were employed in a similar arrangement, a

ground reaction curve was determined for this particular mine

by measuring the load on individual supports and the associated

roof-to-floor convergence in the vicinity of the supports. The

results are shown in figure 10. 

First, it is important to note where these measurements were

obtained. Since abutment loading changes dramatically in

longwall mining as the face approaches, the ground reaction

curve should always be established at the most severe load

condition, which generally will be just behind the tailgate

shield. In this case, the mine wanted to maintain sufficient

control of the tailgate entry to maintain a ventilation airway

back to the next open crosscut. Hence, measurements of

support loading were obtained to distances of 50 to 100 ft inby

the face. For reasons previously explained, a different ground

reaction was measured through the intersections than in the

entries (figure 10). These measurements were also made under

the deepest cover, again to establish the "worst-load" condition.

The following analysis applies to the entries at positions

where there was no influence from the crosscuts. The load on

the four-point wood cribs was estimated at 40 tons with a roofto-floor convergence of 4 in. Qualitatively, it was noted that the

integrity of the immediate roof was showing signs of

deterioration at this convergence, which suggests that 4 in is

approaching critical convergence where failure of the roof is

inevitable. Some of the wood cribs in the area were also

showing signs of premature failure, probably due to poorquality timber. Conversely, convergence in the concrete-cribsupported area was only 0.5 in, and the measured load on the

cribs was 62% greater at 65 tons per crib. These support loads

were converted into a support system load density of 0.625

tons/ft2

 for the wood cribs (equation 1) and 1.35 tons/ft2 for the

concrete stopping block cribs (equation 2).

These two data points were then used to establish the ground

reaction curve for the tailgate entry inby the longwall face as

depicted in figure 10. Since measurements were made on only

these two support systems, a linear approximation to the ground

reaction curve was made using these data as end points with a

straight line connecting them. This curve could then be used to

determine the support load density required to control the

convergence inby the longwall face from 0.5 to 4.0 in. For

example, if the goal were to limit convergence to 2 in, then

drawing a line from 2 in vertically upward until it intersected

the ground reaction curve and then drawing a line horizontally

to the y-axis would reveal that a support load density of 1.04 /ft2

must be provided. An algebraic solution to the problem can

also be found by determining the slope and y-intercept for the

ground reaction curve. Once the algebraic equation for the line

is determined, the support load density at any displacement can

be calculated.

The next requirement is to transform the required support

load density into a support system design. Support load density

is determined primarily by two factors: (1) the stiffness of each

support and (2) the spacing of the supports. Continuing with

our example, if we want to increase the support load density of

the four-point wood crib support system from its current 0.625

to 1.04 tons/ft2

 in order to reduce convergence in the entry from

4 to 2 in, then we would need to decrease the spacing of the

wood cribs from the current 4-ft spacing. The question is by

how much? The required center-to-center spacing to provide a

support load density of 1.04 tons/ft2

 can be determined by first

identifying the load capacity of a wood crib at 2 in of

convergence, which is found from the performance data

developed from the laboratory tests conducted in the Mine Roof

Simulator (figure 2). As shown in figure 2, the capacity of a

four-point wood crib is 27 tons at 2 in of displacement. The

spacing of a single row of cribs is then determined by dividing
141

Spacing(2&in convergence)' Capacity(2in)

Loaddensity× entry width

' 27

1.04 × 16 ' 1.6 ft

(4)

Figure 8.–Arrangement of four-point wood cribs in a western Pennsylvania mine.

this capacity by the product of the required support load density

and entry width. As the following analysis shows, the required

spacing of 4-point wood cribs to achieve a support load density

of 1.04 ton/ft2

 1.6 ft. Such a tight spacing can only be achieved

through a staggered double row arrangement, where the centerto-center spacing in each row is 3.2 ft.
142

Figure 9.–Arrangement of stopping block concrete crib support in a western

Pennsylvania mine.
143

Figure 10.–Measured ground reaction curve for a western Pennsylvania coal mine.

DESIGN METHODOLOGY FOR LONGWALL TAILGATES

The concept examined for the western Pennsylvania mine

can be generalized and employed at any mine to optimize

secondary support design and application. The first step in the

design process is to determine the ground reaction behavior.

This can be done by installing at least two, and preferably three,

support systems of varying support system stiffness and measuring load on and corresponding convergence of the mine roof

and floor. It is important that these supports cover a wide range

of stiffness to provide a full picture of ground reaction behavior. As part of this process, an effort should also be made

to determine the critical roof-to-floor convergence where roof

failure occurs, since this will be a critical design value for the

support design. However, in order to do this, a low density support system must be installed that will allow considerable

convergence to take place. Since this poses a risk of inadequate

ground control, precautions should be taken. One possibility is

to set additional wood cribs through this trial zone with the top

layer removed so that they could be reinstalled to provide additional support if necessary, or in the worst case, provide

ground support after the roof deforms (5-6 in) to the cribs if

they cannot be topped off. 

Altering support system stiffness can be done several ways.

One way is to utilize the same support and same spacing down

the entry, but increase the number of rows of support across the

mine entry from one to two to three, which would proportionally increase system stiffness by the same factors.

Another way is to keep the spacing and number of rows of

support constant, but use supports of varying stiffness. This

would eliminate the impact of both span and roof coverage

which could be limiting factors in support placement (load deformation) strategies. For example, using conventional wood

cribs, the support construction could be varied from a 4-point to

a 9-point to a 16-point crib. The support load density would

increase in direct proportion to the increase in support stiffness.

Adjusting the support spacing down the entry could also be

considered. Support load density is proportionally increased as

the spacing is decreased. Care should be taken to avoid excessive spacing that will cause span-related problems. A good

rule of thumb is that the support spacing should not exceed half

the entry width. The load-bearing area (quality and extent of

the contact area) of the support also an important factor to consider. The pressure exerted by any support should not exceed

the strength of the mine roof or floor. 

Ideally, the loads on the various support systems should be

measured underground. Support load measurements are typically made through a hydraulic flat jack. This will be more

difficult to do on some support designs than others. Supports

with a single contact area, such as concrete cribs, are easier to

work with than something like a conventional wood crib, which

has multiple loading paths. Theoretically, if loads cannot be

measured, they can be estimated directly from laboratory loaddisplacement data provided convergence (support displacement)

measurements are made. However, this estimate may not be accurate for materials such as wood, where creep can occur and

distort the approximation of the load that produced the

measured displacement. Measurements of convergence are
144

GRC safety factor ' Design load density

Minimal acceptable load density (8)

Safety factor (support) ' Peak load capability

Load at installed load density (9)

essential to this design methodology. Roof-to-floor convergence measurements can be made in several ways, but it is

important that these measurements correlate to displacements

induced in the support structure. 

Readings of load versus displacement (convergence) should

be made under the most severe load conditions that occur. For

most longwall mines, this will be at the back of the tailgate

shield. If it is critical that the tailgate be kept open inby the face

for ventilation reasons, then ideally the load-displacement

readings should be made inby the face. Of course, this may not

be as easy to do, but it is important to realize that the design

methodology assumes worst-case load conditions, and safety

factors will need to be employed if measurements are taken

under less severe load conditions. Another good rule of thumb

to follow is that the face should be retreated a distance

approximating the width of the face before the full ground

reaction behavior is established. In other words, if the face

width is 1,000 ft, then the face should be retreated at least 1,000

ft before ground reaction behavior is measured. Of course, the

first requirement for any tailgate design is proper pillar design.

While in theory ground reaction behavior can be determined for

any pillar design and roof geology, the methodology proposed

here assumes that the pillar design falls above the ALPS design

line for a given CMRR [Mark et al. 1994].

The load-displacement data are then used to generate a plot

of support load density as a function of convergence (ground

reaction curve). Each support type with different stiffness represents one data point on the ground reaction curve. The support load density is determined as the measured load in the

support at the observed convergence times the number of

supports per unit area of mine entry (equation 5). For a single

row of supports employed on a constant center-to-center spacing, the support load density can be determined by equation 6.

Support load density ' No. of supports × support load

Area of support coverage (5)

Support load density ' Support load

Center&to¢er spacing × entry width (6)

Once the ground reaction curve is developed, the center-tocenter spacing (down the mine entry) of alternative support

systems arranged in a single row required to achieve ground

control (equilibrium) at a desired convergence can be

determined from equation 7. The center-to-center spacing of a

double row of supports is simply twice that of a single-row

arrangement.

Spacing (displacement) ' Capacity (displacement)

Load density × entry width (7)

Where spacing (displacement) = center-to-center spacing of

a single row of supports in feet,

Capacity (displacement) = Individual support capacity in tons

at a specified displacement (obtained from laboratory performance data) equal to the desired convergence control, tons

Support load density = Support load density in tons/ft2

 at the

required convergence (obtained from ground reaction curve) in

tons per square feet, and

Entry width = width of the entry in feet.

This design methodology is a valuable tool in optimizing the

utilization of standing secondary roof support technology.

However, as previously described, it is still up to the mining

engineer to decide how close to the critical convergence he/she

wants to operate based on knowledge of the particular ground

conditions. A margin of safety is provided by designing for a

convergence that is less than the critical convergence (minimal

acceptable support load density). To make equivalent comparisons of alternative support systems, a safety factor can be

quantified by comparing the design support load density to the

minimal acceptable support load density that will be representative of the maximum allowable (critical) convergence.

This is referred to as the ground reaction curve (GRC) safety

factor (equation 8).

Another factor to consider is whether the support is being

fully loaded and how much reserve capacity is left in the

support at the design load. In the event that load conditions

worsen beyond expectations, this reserve support capacity may

be needed to support the mine roof. If the support characteristic

is such that the support sheds load quickly after reaching its

peak load, such as the nonyielding concrete supports (figure 5),

then consideration must be given to avoid designing near the

peak loading capability of the support. A safety factor for the

support can be defined based on support loading at the required

support load density in relation to the peak loading capability of

the support (equation 9). Hence a safety factor of 1 indicates

that there is no reserve capacity available, and a safety factor of

2 indicates that the support is loaded to only 50% of its maximum support capacity.
145

ALTERNATIVE SUPPORT STUDIES AT THE WESTERN PENNSYLVANIA MINE

Four different standing support systems and one cribless

system were installed in the longwall tailgate entry of this mine.

In addition to conventional four-point wood crib supports and

concrete stopping block supports, the alternative standing supports were (1) Heintzmann Corp.'s Alternative Crib Supports

(ACS’s), (2) HeiTech Corporation’s Pumpable Crib Supports,

(3) Strata Product’s Propsetters, and (4) Burrell Mining Products'

The Can. Cable trusses were used in the cribless area. The alternative supports were assigned to sections of the tailgate between

the standard wood cribbing. This was done to ensure that there

was no interaction between support installations, thereby allowing

a fair evaluation to be made under equivalent conditions (i.e., normal cover, no excessive roof or floor damage, normal geology).

Table 1 shows the installed spacing of the alternative support

systems and the typical 4-point wood crib system and concrete

stopping block crib system. The support load density of the alternative supports is calculated by matching system performance

to the ground reaction curve. Essentially, this requires working

backward through the design methodology. The following steps

can be used.

1. Pick an arbitrary convergence within the bounds of the

ground reaction curve.

2. Determine the required support load density that matches the

ground reaction curve for this convergence.

3. Identify the individual support capacity at this displacement

from the laboratory performance data at this convergence.

4. Determine the support load density from equation 6.

5. If the support load density is greater (falls above the curve)

than the required support load density, a lower convergence

should be chosen. If the support load density is less (falls

below the curve) than the required support load density, then

higher convergence, should be chosen and steps 1 through 4

repeated until the support load density matches the ground

reaction curve.

An analysis of table 1 reveals that all four alternative support

systems were installed with sufficient support load density to

control the convergence well below the critical level of 4.0 in

provided by the 4-point wood cribs. The high safety factor

utilized in these alternative support applications was to provide

a margin of safety in anticipation of a tailgate horizontal stress

concentration on the next panel.

The HeiTech pumpable support had the highest load

density at 1.35 tons/ft2

, which limited convergence to approximately 0.5 in. Conditions both inby and outby the face

as shown in figure 11 were excellent with the HeiTech pumpable support system. However, it should be noted that the

HeiTech support also had the lowest support safety factor

(1.3) of the four alternative support systems utilized, meaning

that load development approached the peak loading capability

of the support. If the maximum loading capability of the support was exceeded due to unexpected additional roof loading

or variability in the peak strength of the support, convergence

would increase to approximately 4 in at the installed spacing

based on the residual support capacity of approximately

90 tons. Since this is the critical convergence for this mine (at

which point roof conditions deteriorate significantly), it is

critical that the peak pumpable support capacity not be exceeded through the zone where it is desired to maintain full

roof control, which means that the spacing must be properly

maintained during installation so as not to overload the

support past failure. 

Table 1. Assessment of standing alternative support technologies utilized in study at a western Pennsylvania. mine.

Support system Installed

spacing, ft

Installed load

density, tons/ft2

Conv. control, 

in

Safety factors Observed roof condition

GRC1 Support2 Outby face Inby face

Four-point wood crib . . . . . . . . . . . . 8.0 (DR3

) 0.625 4.0 1.00 1.8 Good Marginal

Concrete stopping block crib . . . . . . 3.0 (DR) 1.35 0.5 2.16 (04

) 1.0 Good Marginal

Heintzmann ACS . . . . . . . . . . . . . . 5.0 (DR) 1.20 1.24 1.92 2.5 Excellent Good

HeiTech Pumpable concrete crib . . 9.2 (SR) 1.35 0.5 2.16 1.3 Excellent Excellent

Strata Products Propsetter . . . . . . . 4.0 (DR) 1.12 1.6 1.79 1.7 Excellent Good

Burrell Mining Products Can support 7.0 (DR) 1.19 1.25 1.90 1.8 Excellent Excellent

1

Ground reaction curve safety factor is determined from equation 8 as the ratio of installed support load density to minimum allowable support load

density.

2

Support safety factor is determined from equation 9 as the ratio of peak loading capability of the support to load developed at installed spacing. 3

All double rows of supports were installed in a staggered fashion. The spacing here refers to the spacing of one row of supports. With the

staggered arrangement, the spacing between adjacent supports of both rows is half of that of the individual row (see figure 8 or 9). 4

The roof condition was good prior to failure of the support. Hence, the installed support load density actually dropped to zero once the support

failed, which accounted for deterioration in the integrity of the roof.

DR = Double row. SR = Single row.
146

Figure 11a.–Outby area supported with pumpable crib.

Figure 11b.–Inby area supported by pumpable crib.

Figure 12a.–Outby area supported by Propsetter support.

The Propsetter support had the lowest margin of GRC safety

at 1.79, but even this system was conservative in that the convergence was limited to 1.6 in. Conditions outby the face were

excellent, as shown in figure 12A, and relatively good inby the

face (figure 12B). Some of the Propsetter supports well inby

the face (mostly in the intersection areas) (figure 12C) appeared

to be in a state of post-yield deformation where “brushing”

(yielding) caused the props to tilt from a vertical orientation,

which is normal for this load condition. It does not mean that

the prop is shedding load. Another possibility is that the props

were being dislodged or moved laterally by flushing of the gob

material, floor heave, and/or lateral displacements of the roof

relative to the floor by the cantilevered roof beam. Despite

these occasional abnormalities in the support condition, the

Propsetter was able to maintain an effective air way beyond the

first open crosscut inby the face. It was also reported by the

mine that five or six Propsetters were dislodged from the mine

roof and floor outby the face. Since convergence was minimal

outby the face, the cause of these props “falling over” was

never definitively determined. The same props were reinstalled

and performed well throughout the duration of the test.

The Heintzmann ACS support had the most limited yield

capability of the four alternative support technologies used at

this mine. The ACS also shed load rather quickly after reaching

its peak loading capability at about 2.2 in (figure 7). However,

the installed spacing provided the highest support safety factor

(2.5), meaning the loads were kept well below the peak capacity

of the support. Likewise the installed load density limited the

convergence to 1.2 in, which is considerably less than the yield

point of 2.2 in. Hence, this is a good example of how a stiff

support with limited yield capability can provide effective

ground control in a longwall tailgate, provided that a sufficient

number of supports are installed per unit area to establish a high

enough load density to minimize the ground movement.

Figure 13 shows the condition of the entry both outby and inby

the face in the area supported by ACSs. Similar to the Propsetter support, a few of the ACS props were tilted inby the face,

but continued to provide support capability in this condition

without becoming unstable.

The Burrell Can support installation had a GRC safety factor

(1.92) almost identical to that of the area supported by the

Heitzmann ACS. The 1.92 GRC safety factor means that the

installed load density was almost twice that needed to prevent

roof failures from occurring. The entry conditions, both inby

and outby the face, were excellent with the Burrell Can support,

as shown in figure 14. Inby, the conditions were slightly better

than in the area supported by the ACS. This improvement is attributable to the larger surface coverage and improved stability

of the Burrell Can support compared to the ACS.
147

Figure 13a.–Outby area supported by ACS supports.

Figure 13b.–Inby area supported by ACS supports.

Figure 12b.–Inby area supported by Propsetter support.

Figure 12c.–First open crosscut intersection inby the face

supported with Propsetter supports.
148

Figure 14a.–Outby area supported by Burrell Can support. Figure 14b.–Inby area supported by the Burrell Can support.

USING THE GROUND REACTION CURVE TO OPTIMIZE THE USE 

OF ALTERNATIVE SUPPORTS

Once the ground reaction curve is determined, use of any other

support technology can be defined by strategic employment strategies relative to the ground reaction curve. Table 2 shows alternative placement strategies for these and other alternative support

systems for this particular mine site based on measured ground

reaction behavior with conventional wood and concrete support

systems.

Examining table 2 reveals that other support technologies

could be used favorably. One example is the Link-N-Lock crib

support developed by Strata Products. A Link-N-Lock crib

24 in long could be installed in a staggered double row with a

center-to-center spacing of 11 ft per row (5.5 ft between cribs

in opposite rows); this support could limit convergence to 2 in.

Another alternative would be a single row of 36-in Link-NLock cribs on an 8-foot center-to-center spacing, which also

would limit convergence to 2 in. In contrast, a single row of

nine-point cribs would have to be installed on a 5.2-ft center-tocenter spacing to provide equivalent ground control capability.

Three other points can be made by examining the data in

table 2. First, the spacing of stiff, high-capacity support systems

can become excessive at large displacements. It is important to

remember that ground reaction behavior was measured in the

immediate vicinity of secondary support. It is assumed that

support loading is sufficiently transferred to control the roof and

floor between the supports. Obviously, there are some limitations

to this capability, which largely depend on the strength of the

immediate roof. Generally, stronger roofs can span greater distances between supports than weaker roof. Currently, this

capability is best obtained from empirical data within the mine,

but a good rule of thumb to follow in the absence of specific

information is that the span between supports should not exceed

half the entry width, particularly in weak roof conditions such as

those observed in the Pittsburgh coal seam. Using this criterion,

it is seen from table 2 that nine-point cribs and Link-N-Lock cribs

must be employed at a load density greater than 0.63 tons/ft2

 to

avoid an excessive spacing where failure might occur between the

cribs. Surface control of the immediate roof is another issue.

Surface control refers to failure of the immediate skin of the roof.

This is different from the excessive spacing issue discussed above

in that there is no major failure of the roof rock. If surface control

is necessary to prevent flaking of the roof skin between cribs, then

methods such as wire meshing can be effectively employed as a

control measure. 

Second, it is seen from table 2 that stiff concrete supports

must be installed at a high enough support load density to limit

convergence to less than 2 in; otherwise, the supports will fail

prematurely, resulting in no support capability and unstable

ground conditions after failure of the support. Third, it is seen

that required load density to limit convergence below 1 in is not

practical with passive wood crib supports, including the stiffer

Link-N-Lock cribs. Wood is just too soft to generate meaningful loads at such small displacements.
149

Table 2. Recommended placement and safety factors for alternative support technologies.

Support system

Center-to-center support spacing and individual support capacities at convergences of 0.5,1, 2, and 4 in.

0.5 In

(LD =1.35), (SF =2.16)

1 IN

(LD = 1.24), (SF = 2.0)

2 In

(LD = 1.04), (SF = 1.7)

4 In

(LD = 0.63), (SF = 1.0)

Load (tons) Space (ft) Load (tons) Space (ft) Load (tons) Space (ft) Load

(tons)

Space (ft)

Four-Point cribs . . . . . . . . . 8 0.7 17 0.9 27 1.6 39 3.9

Nine-Point cribs . . . . . . . . . 14 0.6 55 2.8 86 5.2 115 11.4

24-in Link-N-Lock . . . . . . . 10 0.5 45 2.3 92 5.5 115 11.4

27-in Link-N-Lock . . . . . . . 11 0.5 49 2.5 102 6.1 127 12.6

36-in Link-N-Lock . . . . . . . 13 0.6 63.5 3.2 132 8.0 162 16.1

Propsetter (8.5 in dia) . . . . 12 0.6 25 1.3 42 2.5 57 5.7

Stopping block cribs . . . . . 65 3.0 0 N/A 0 N/A 0 N/A

SFR donut cribs . . . . . . . . 86 4.0 280 14.1 0 N/A 0 N/A

SFR block (2 per layer) crib 85 4.0 210 10.6 0 N/A 0 N/A

HeiTech Pumpable Crib . . 190 8.8 240 12 112 6.7 90 8.9

Burrell Can (24-in dia) . . . . 40 1.9 65 3.3 90 5.4 90 8.9

Heintzmann ACS (100 ton) 39 1.8 46 2.3 102 6.1 36 3.6

N/A - Indicates that support would fail prior to the designated convergence and would not have sufficient post-failure (residual) capacity to be

considered for use in this condition. LD designates the support load density of a single row of cribs in the designated center-to-center spacing.

SF refers to the GRC safety factor as computed by equation 8 for a single row of cribs. If two rows of supports were used, the designated spacing

at a specific displacement would be reduced by a factor of 2, and the safety factor would be increased by a factor of 2.

OTHER CONSIDERATIONS IN TAILGATE SUPPORT SELECTION AND APPLICATION

OF THE PROPOSED DESIGN METHODOLOGY

While the primary consideration in support design is obviously

the prevention of roof falls through proper ground control, there

are other factors to consider. These include (1) cost of the support, (2) material handling requirements and ease of installation,

and (3) impact of the support structure on ventilation. These

issues are beyond the scope of this paper. 

While this paper is focused on standing roof support applications, several mines have explored the application of intrinsic secondary support, such as trusses, to replace conventional

standing support in longwall tailgates [Mucho 1998]. Two

points need to be made in reference to truss supports. First, it

should be noted that the design methodology proposed in this

paper applies only to standing roof support. While some of the

basic rock mechanics principles used here may apply to intrinsic

support, the support mechanisms are different, and these have

not been examined in this study.

Another caveat of the design methodology pertains to

application in yield pillar gate roads. While in theory a

ground reaction behavior can be established for yield pillar

systems, the mechanisms of ground behavior and support

interaction are different. In particular, the yield pillar system

is a high deformation environment by design. Secondary

support should ideally allow the ground to yield in accordance

with pillar deformation and not interfere to the point where the

secondary support develops sufficient capacity to damage the

roof while it is yielding. Hence, a stiff, high-density support

may not be desirable in this environment, and the important

secondary support design consideration may well be the

stability and yield capability of the support.

CONCLUSIONS

Several alternatives to conventional wood and concrete

cribbing have been developed in recent years. These new

support technologies provide improvements in supporting

capability as well as material handling advantages. However,

since their supporting characteristics are all different, a design

methodology must be developed so that for mines can employ

these technologies and safely maximize their benefits without

increasing the overall cost of support. Conservative applications or trial-and-error assessments are no longer practical nor

prudent for state-of-the-art longwall mines.

The design methodology proposed in this paper and examined through a field trial at a western Pennsylvania coal mine

embodies a fundamental concept of rock mechanics, that being

the "ground reaction curve." Measurement of the ground reaction curve at this mine indicated that support capacity had a significant impact on the ground behavior in the longwall tailgate.

Increasing the support load density by a factor of 2 from 0.625

to 1.25 tons/ft2

 decreased convergence in the entry from 4 to 1

in. Conventional four-point wood cribs installed in a double

row with an 8-ft center-to-center spacing in a staggered
150

arrangement (4-ft center-to-center spacing between adjacent

rows) resulted in marginal ground control. Concrete cribs constructed from concrete stopping blocks reduced convergence to

0.5 in, but some failed under this amount of deformation, resulting in localized poor ground control resulting from support

failure. 

Four alternative standing support technologies were installed

at the western Pennsylvania mine: (1) Can (Burrell Mining

Products), (2) Alternative Crib Support (Heintzmann Corp.), (3)

Pumpable concrete support (HeiTech Corp.), and (4) the

Propsetter (Strata Products USA). These alternative support

technologies were installed at a support load density ranging

from 1.12 to 1.35 tons/ft2

, providing ground control safety factors of 1.79 to 2.16. Ground conditions for all these support

applications were generally very good, which is consistent with

the measured ground reaction behavior and installed support

density.

The NIOSH Support Technology Optimization Program

(STOP) has been developed to facilitate the use of this design

methodology and allow mines to optimize the use of any support technology once a ground reaction curve for that particular

mine has been identified [Barczak 2000].

REFERENCES

Barczak TM [1994]. Assesment of wood and alternative materials for

secondary roof support construction. In: Peng SS, ed. Proceedings of the 13th

International Conference on Ground Control in Mining. Morgantown, WV:

University of West Virginia, pp. 190-201.

Barczak TM [2000]. Optimizing secondary roof support with the NIOSH

Support Technology Optimization Program (STOP). In: New Technology for

Coal Mine Roof Support. Pittsburgh, PA: U.S. Department of Health and

Human Services, Public Health Service, Centers for Disease Control and

Prevention, National Institute for Occupational Safety and Health, DHHS

(NIOSH) Publication No. 2000-151, IC 9453.

Deere DU, Peck RB, Monsees JE, Schmidt B [1970]. Design of tunnel

lining and support systems. Highway Research Record, Washington, DC, No.

339, pp. 26-33.

Mark C, Chase FE, Molinda GM. [1994] Design of longwall gate entry

systems using roof classification. New Technology for Longwall Ground

Control. Pittsburgh, PA: U.S. Department of the Interior, Bureau of Mines,

Special Publication 01-94, pp. 5-17.

Mucho TP [1998]. In focus—cable bolts: a “new” support. Falls Church,

VA: U.S. Department of Labor, Mine Safety and Health Administration,

Holmes Safety Association Bulletin Mar:3-4.
151

OPTIMIZING SECONDARY ROOF SUPPORT WITH THE NIOSH

SUPPORT TECHNOLOGY OPTIMIZATION PROGRAM (STOP)

By Thomas M. Barczak1

ABSTRACT

The 1990s brought an unprecedented increase in the development of innovative technologies to provide

more effective and easier-to-install roof support in underground mines. To facilitate the application of these

technologies in improving mine safety, the National Institute for Occupational Safety and Health (NIOSH)

developed the Support Technology Optimization Program (STOP). STOP is a Windows-based software

program that provides mine operators with a simple and practical tool to make engineering decisions about the

selection and placement of these various roof support technologies. The program includes a complete database

of the support characteristics and loading profiles obtained through safety performance testing of these

supports at the NIOSH Safety Structures Testing Laboratory. A support design criterion in the form of the

required support load density at a specified convergence can be established from four options: (1) a database

of measured ground reaction obtained from various mines or ground behavior information input by the user,

(2) load requirements based on a detached roof block or rock failure height, (3) criteria based on the current

roof support system, and/or (4) arbitrary criteria set by the user. Using these design criteria, the program will

determine the installation requirements for a particular support technology that will provide the necessary

support load density and convergence control. Optimization routines are also available to determine the most

efficient support design for a user-specified support installation. In addition to these performance measures,

STOP can be used to compare material handling requirements and installation costs. Comparisons among the

various support technologies are easily made, including a graphical analysis of relevant support parameters.

This paper describes STOP and its application to optimizing standing secondary roof support systems.

1

Research physicist, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
152

INTRODUCTION

Secondary support provides additional roof support in the

event of failure of the primary support system. When properly

designed, secondary support will also assist the primary support

in controlling the integrity of the immediate roof beam. Thus,

roof support should be thought of as a three-part system:

(1) the remaining coal pillars, which provide control of

overburden weight, (2) the primary support system consisting

of roof bolts, which help form a more competent roof beam

and, in the case of mechanical bolts, attach the immediate roof

beam in suspension to the more stable main roof rock, and

(3) the secondary roof support system, which consists of

standing roof support and intrinsic support elements designed

to control deformation of the immediate roof and handle

additional abutment loads during retreat mining. The latter

occurs in longwall mining where the tailgate is frequently

supported with various secondary roof support systems. It

should be noted that the Support Technology Optimization

Program (STOP) in its present form is limited to the evaluation

of standing roof support systems.

Engineering design is applied primarily to size the pillars to

account for load variations due to depth of cover, active mining

zones, and the quality of the roof rock. Computer programs

such as the Analysis of Longwall Pillar Stability (ALPS) and

Analysis of Retreat Mining Pillar Stability (ARMPS) programs

are valuable tools in designing pillars for various mining scenarios [Mark 1992; Mark and Chase 1997]. There are no universally accepted design criteria for primary support (roof

bolts), although recent research by the National Institute for

Occupational Safety and Health (NIOSH) indicates that there

may be some fundamental criteria to define conditions where

current primary support densities are inadequate. The Wood

Crib Performance Model [Barczak and Gearhart 1993] was developed by the former U.S. Bureau of Mines in 1994 to provide

engineering design for conventional wood cribs, which were the

predominant form of secondary support at the time. The Wood

Crib Performance Model was used to determine the supporting

capability of various conventional wood crib configurations and

match this capability to user-defined load and convergence criteria. More recently, a design methodology for standing secondary roof support in longwall tailgates was developed that

incorporates measured ground reaction data into the formulation

of the load and convergence design criteria for standing roof

support systems [Mucho et al. 1999; Barczak et al. 1999].

In the past 5 years, there have been 16 new standing roof

support systems developed for use as secondary roof support.

These new support systems not only provide superior roof support, but many provide material handling advantages as well.

STOP was developed by NIOSH to allow mine operators to

compare these various support systems and to optimize the application of both new standing roof support technologies and

conventional wood and concrete crib support systems.

Although STOP can be considered as an upgrade of the original

Wood Crib Performance Model, it is built on a Windows-based

architecture and has several enhanced features that were not

available in the previous Wood Crib Performance Model.

These includeC

(1) Selection from a database of currently available standing

roof support systems for evaluation;

(2) A synopsis of pertinent design and installation criteria for

each support system;

(3) A description of performance characteristics, including

photographs of the support loading profile showing the condition of the support as it deforms;

(4) Name and telephone numbers for support manufacturers;

(5) Ground reaction curve support design criteria where the

laboratory support performance can be matched to a curve corresponding to ground behavior, as opposed to a single (load and

convergence) data point, as was done in the original Wood Crib

Performance Model;

(6) Enhanced optimization algorithms that determine the

most sufficient support design for user-specified spacing limitations and/or the user-defined load density and convergence

requirements; 

(7) Material handling and cost information for each support;

and

(8) graphical displays of support system capabilities.

It is important to understand that although there are now a

wide variety of support choices, each of these support systems

has a unique performance profile. Simply replacing one

support system with another will not provide equivalent ground

control. Most of the new support technologies provide superior

supporting capability, which may allow wider spacings of the

support to be used if the goal is to provide support capability

equivalent to that of a conventional wood crib support system.

STOP will determine the spacing requirements that will provide

equivalent support capacity. This is one way that STOP can

optimize the use of a particular support system. STOP can also

provide important information regarding the benefits of increasing support load density. Using measured ground reaction

data, STOP will determine either the convergence that can be

expected from a certain support design installed on userselected spacings or the support spacing required to limit

convergence to a certain level.

This information can be very useful when petitioning the

Mine Safety and Health Administration (MSHA) for approval

to use an alternative support technology. Without this information, MSHA will typically require a trial section where the

alternative support system can be observed before full approval

is granted. In longwall mining, this means that a mine operator

might have to wait for a full panel of mining before
153

implementing a new support technology. Likewise, without an

engineering basis to justify a change, variations in placement

strategy or implementing a change in support design can be

delayed until a trial section is observed. Thus, STOP can be included as another part of the overall process that MSHA may

use in approving a roof support plan. While it may not be the

sole deciding factor, STOP can provide critical engineering data

that will facilitate a decision regarding the implementation of

these new support technologies.

This paper introduces STOP, describes the architecture of

the program, and provides several examples of how it can be

used to optimize the design and use of secondary roof support

technologies. STOP can provide an engineering foundation to

ensure that inadequate support designs, as well as ultraconservative support applications, are avoided. Safety will be

improved by proper matching of support performance to mine

conditions, which will reduce the likelihood of roof falls and

blocked escapeways. Material handling injuries associated with

support construction are known to account for about 5,000 lost

workdays per year in underground coal mines. STOP can help to

define the material handling advantages of alternative support

technologies that use lighter weight materials or systems that can

be installed with mechanical assist. The use of these support

technologies can significantly reduce material handling injuries.

PROGRAM ARCHITECTURE

STOP is a Windows-based architecture. The Main Menu

allows the user to control the flow through the program if

desired. This window can be accessed through each of the

primary program segments. The Main Menu contains six

modules: File, Design Criteria, Support Evaluation, Comparison, Information, and Help.

File: The File module contains file management

subroutines that allow the user to create new files, open and

close existing files, and exit from the program. The File menu

also allows the user to set the path for storing several

photographs that are incorporated into the program to allow

visual display of support performance.

Design Criteria: The Design Criteria module is where the

load and convergence design criteria for the support system are

formulated. The requirement is to define the required support

load density in terms of tons of support capacity per linear foot

of entry advance and at what convergence this support

capability is to be provided. There are four different ways to

establish these design criteria in the program: (1) ground reaction curve, (2) detached block, (3) current support system, and

(4) arbitrary criteria.

1. Ground Reaction Curve (figure 1) allows the user to

define support load density and convergence criteria from inmine measurements of the ground behavior (convergence) associated with various support systems [Mucho et al. 1999;

Barczak et al. 1999]. Essentially, the ground reaction concept

implies that convergence in the mine entry is controlled by the

magnitude of support resistance. Generally, convergence decreases with increasing support load density. Thus, if measurements of convergence are made with two or more support systems of varying stiffness, then a ground reaction curve can be

established for that particular mine. The user can define a

ground reaction curve or use one from the database established

from various mine sites maintained in the program. Once a

ground reaction curve is defined, the program will determine

the required spacing for a particular support system that will

provide the support load density consistent with the ground

reaction curve at a specified convergence.

2. Detached Block is shown in figure 2. The support load

density requirements are established by calculating the weight

of a detached block of roof rock above the mine entry. The

failure height can be inputted by the user or estimated from the

quality of the roof rock (Coal Mine Roof Rating) using an

approximation developed by Unal [1986]. The volume of the

block is also influenced by the shape of the failure. Options

include either an arch or a vertical shear failure at the pillar

boundaries. Two options are available for determining the convergence criteria. If ground reaction information is available,

this information can be used to help define the convergence

criteria. In terms of the ground reaction curve, there is a critical

convergence where failure of the roof occurs. This could be

used to define the convergence criteria for the detached block,

the idea being that the support should put the roof rock mass

into equilibrium before the critical convergence is reached. If

this option is selected, convergence is defaulted to the maximum convergence on the ground reaction curve, but the user

can change this input if desired. In the absence of ground reaction information, the user can simply input a convergence

criteria (allowable displacement before roof weight is supported

in equilibrium).

3. Current Support System allows the user to define design

requirements based on the performance of the current support

system (figure 3). Two options are available. The first one is

that if a ground reaction curve is available, then the program

will determine where the current support system falls on the

ground reaction curve and set the support load density and

convergence requirements to that point. For the second option,

the user must define an allowable convergence, which should

be based on some in-mine measurements. The support capacity

and resulting load density for support spacing will be calculated

from the load-displacement profile for the support, as
154

Figure 1.—Window for establishing design criteria based on ground reaction curve.

determined from tests in the Pittsburgh Research Laboratory's

(PRL) unique Mine Roof Simulator load frame or from other

laboratory data.

4. Arbitrary Criteria simply allows the user to set the

support load density and convergence criteria to any arbitrary

set of values.

Support Evaluation: The Support Evaluation module is the

heart of the support design process. Any of a variety of support

systems contained in the program database can be selected for

design and analysis (figure 4). The Design algorithm calls up

a subroutine that allows the user to control relevant design

parameters and/or pick from the available design (models) for

a particular support type (figure 5). The user must also input

the number of rows to be used in the support placement. The

program will then calculate the required spacing of the supports

to achieve the desired support load density at the designated

convergence, or the user can select a support spacing and the

achieved convergence will then be calculated for the userdefined support installation. An optimization algorithm is also

included in which the program will determine the support

design or model that most closely matches the design criteria

(support load density at designated convergence). Also included in the Support Evaluation module are analyses of

installed support costs and material handling requirements for

the support.

Comparison: The Comparison module compares the support systems chosen for analysis. There are three windows in

the Comparison module: (1) Comparison Assessment Table,

(2) Support Description Summary, and (3) Graphical Data

Analysis. The Comparison Assessment Table describes the support layout (number of rows and support spacing) and various

design parameters for each support system in a tabular format.

The parameters are grouped into six categories: (1) support

layout, (2) ground control, (3) unit support costs, (4) normalized

support cost, (5) installation assessment, and (6) material

handling. The user can pick any one of the selected support

systems as a baseline system for comparison purposes. The

Support Description Summary summarizes the support design

parameters for each support system. The Graphical Data

Analysis window allows the user to plot support performance

(unit support load or support load density) as a function of convergence and graphically compare the various support systems,

as shown in the example in figure 6. Ground reaction data can

also be displayed on the plots to show the convergence control

provided by the various support systems relative to the ground

reaction curve.

Information: The Information module can be thought of as

a general information center. The various support technologies

are categorized in six groups: (1) conventional wood (crib)

supports, (2) engineered timber supports, (3) conventional
155

Figure 2.—Window for establishing design criteria based on detached roof block.

Figure 3.—Window for establishing design criteria based on current roof support system.
156

Figure 4.—Window for selecting support types for design and analysis.

Figure 5.—Window for designing support system and determining installation requirements to achieve

design criteria objectives.
157

Figure 6.—Graphical analysis of selected support systems showing support load density as a function

of convergence.

concrete crib supports, (4) yieldable concrete supports, (5) steel

supports, and (6) additional supports. From this list and the

embedded submenus, the user can select a specific support and

learn more about the support through several other program

buttons. Design Information provides a description of the support, design and installation considerations, performance

characteristics, and manufacturer or supplier contact information. Performance displays the loading profile of the support

with photographs that depict deformation and associated

support loading. NIOSH Testing Laboratory describes PRL's

Mine Roof Simulator and refers to the safety performance

testing protocols through which the performance characteristics

of the support were determined. Reference/Bibliography contains relevant reference material pertaining to the selected

support system.

Help: A context-sensitive Help file is available to facilitate

operation of the program and interpretation of the results. The

Help file can be called from each window or from the main

menu.

HOW TO USE STOP

Generally, the program control guides the user through a

logical sequence of operations to facilitate the design and

implementation of a roof support system (figure 7). A General

Program Flow window is shown on startup. This window

shows the basic program flow and recommends using the Next

buttons to assist the user in following this recommended

procedure for support design and analysis. The Next button

transfers control to the Design Criteria module, where the user

must select the basis for establishing the design criteria by

choosing one of the following options: (1) ground reaction

curve, (2) detached block, (3) current support system, or

(4) arbitrary criteria. Control is then transferred to the

appropriate window for the chosen design criteria and the user

then defines the support load density and convergence design

criteria in that window.

Once the design criteria are established, control is

transferred to the Select Supports window. Here the user picks

the supports to evaluate. Several options are available: (1) Add

allows the user to select a new support and review the NIOSH

database on support performance and design considerations,

(2) Delete deletes a support from consideration, (3) Duplicate

duplicates the choice of a support, which can be helpful when

the user wants to reevaluate a support design with a few minor

changes, and (4) Rename simply renames the support.
158

Figure 7.—Flow diagram showing program control.

After a support type is selected, control is transferred to a

window where the user can define the appropriate support

design parameters. Once the user defines the support design,

the program calculates two outputs: (1) Achieved Ground Control, where support load density and the convergence control

provided by the support system are displayed, and

(2) Installation Requirements, where the number of rows and

required support spacing are provided. For most supports, under Support Specifications, there is also an Optimization button.

Clicking on the Optimization button causes the program to

transfer to an optimization window where the user can select a

support spacing and number of rows and the program will

calculate the support model or design that most closely matches

the required convergence and support load density previously

established in the design criteria. The output of the optimization algorithm depends on which spacing option was selected

in the support specifications window. If the "Calculate the Required Spacing" option is selected, the optimization routine will

select a support design that will meet the design criteria at less

than the specified maximum spacing. If the "User-Defined

Spacing" option is selected, the program will determine a

support design that will meet the load density requirement at a

convergence less than the design convergence.

Once the support system is defined and the installation requirements (number of rows and spacing) are determined,

control is transferred to the materials handling window where

the support costs and material handling requirements can be

defined and examined. A set of default values are included in

the program that are considered to be representative of the

various support technologies at the present time; however, the

program allows the user to modify any of these parameters. In

particular, the cost parameters may be mine-specific and timedependent to some degree. These default values will be

updated periodically when STOP is eventually placed on the

NIOSH Web site (www.cdc.gov/niosh); however, the user

should contact the support manufacturers to receive the latest

cost information. Finally, the Next button transfers control to

the Comparison module, where the various support systems can

be compared to one another.
159

Figure 8.—Example of load profile display available in Information Center module.

EXAMPLES OF APPLICATIONS OF STOP

EXAMINING THE LOADING PROFILE

OF A SUPPORT SYSTEM

In the Information Center of STOP, photographs of each

support during its loading phases are shown when the Performance button is clicked. The load-displacement curve for

the support is shown with a vertical line to designate the

displacement that corresponds to the photo in the window.

Photos are typically shown at 2-in increments in support

displacement. An example for the Propsetter support is shown

in figure 8.

OPTIMIZING THE USE OF CONVENTIONAL

WOOD CRIBS

Historically, conventional wood cribs have been used

extensively for secondary roof support. A common support

system is a double row of 4-point cribs constructed from 5- by

6-in (cross-sectional dimensions), 30-in-long, mixed hardwood

timbers. STOP can be used to evaluate alternative designs and

show that 9-point cribbing can be more cost-effective. The

procedure to conduct such an analysis would be as follows:

1. Choose Current Support Systems as the basis for

selecting the design criteria. Change entry height to 84 in for

this example. Since no supports have been defined yet, click on

the Add Option to transfer control to the Add Supports to

Analysis window, where wood cribs can be selected and

relevant information on the design and performance of wood

cribs can be reviewed. OK then transfers control to the Wood

Crib Specifications window.

2. In the Wood Crib Specifications window, enter the wood

crib specifications (timber width, timber thickness, timber

length) and select the mixed hardwood species to establish

wood strength. Input the number of timbers per layer. After

confirming the support design, OK transfers control to the

Select Current Support window. OK then transfers control to

the Design Criteria window, where the current wood crib

design is featured.

3. Enter a value for the spacing of the supports, number of

rows, and a convergence to establish the support load density

design criteria. In this example, a spacing of 81 in for a double

row of cribs and a convergence of 4 in were chosen. OK transfers

control to the Select Basis for Design Criteria window, where the

support load density requirements of 10.6 tons/ft and

convergence control of 4 in are displayed. When these values are

confirmed by pressing Next, the Select Support window is

recalled. The user is required to update this design by activating

the Design and Cost button before proceeding. When the Design

and Cost button is pressed, control transfers to the Wood Crib

Performance window, where the Installation Requirements and

Achieved Ground Control Parameters are displayed.
160

Figure 9.—Comparison of installation costs of various wood

crib designs (timbers are 5×6×30 in and placed with 5-in side

down).

Figure 10.—Same analysis as shown in figure 9 except

wide (6-in) side of timber placed down.

Figure 11.—Evaluation of engineered timber support

systems as a replacement for conventional wood cribbing.

4. The support material handling and costs for this crib design are then examined by pressing the OK button.

5. When control transfers back to the Select Supports window, the 9-point crib can be analyzed. One way to accomplish

this is to duplicate the current support (press Duplicate button

and then the Design and Cost button) and simply change the

number of timbers per layer from 2 to 3 and the number of rows

from 2 to 1. The program will then calculate the required spacing of single row of 9-point cribs that will provide the same

load density as that of the double row of 4-point cribs.

Figure 9 documents the result of one such analysis and

shows a comparison of the installed cost of a double row of

4-point cribs on an 81-in center-to-center spacing with that of

single row of 9-point cribs on a 92-in spacing. Both support

systems, using cribs constructed from 5×6×30-in mixed hardwood timbers, provide 10.6 tons/ft of support capacity at 4 in of

convergence. Also included in this analysis is a double row of

4-point cribs constructed from all oak timbers instead of mixed

hardwoods. Note that in this analysis, the narrow (5-in) side of

the timber was placed down to establish the interlayer contact.

Figure 10 illustrates the same comparison, except that the

cribs are constructed with the wide (6-in) side down instead of

the narrow (5-in) side down, as was done in the previous

example. The results clearly show the benefits of maximizing

support capacity by increasing the contact area using the wideside-down construction.

REPLACING CONVENTIONAL WOOD CRIBBING

WITH ENGINEERED TIMBER SUPPORTS

In recent years, numerous alternative timber supports have

been developed. These supports are engineered to provide improved loading characteristics compared to conventional wood

cribbing. For this example, the goal is to replace a conventional

wood crib design with engineered timber supports and provide

equivalent support capability in terms of support load density

at a specific convergence. The procedure for designing these

engineered timber supports is essentially the same as in the previous example, except that alternative supports are chosen for

analysis instead of conventional wood cribs.

The baseline case for this example is the same as that chosen

in the previous example: a double row of 4-point, mixed hardwood cribs constructed from 6×6×36-in timbers on 116-in spacing providing 10.52 tons/ft support capacity at 4 in of convergence. The alternative supports chosen for this example were

(1) 24-in Link-N-Lock, (2) 30-in Link-N-Lock, (3) Hercules

HM-9(308) crib, and (4) 30-in Tri-Log crib. Figure 11 shows the

installed support cost per foot of entry for support systems

designed by STOP to provide equivalent support loading to that

of the conventional wood crib support system chosen as a baseline. The installation requirements are also shown. As seen in

figure 11, all four of the engineered timber support systems are

able to reduce the installed support cost considerably without

sacrificing support capability.

INCREASING SUPPORT LOAD DENSITY TO

REDUCE ENTRY CONVERGENCE

The objective of increasing support load density is to

improve ground control by allowing less roof movement. If the

ground reaction at a particular mine is known, then support

systems can be designed to provide any measure of convergence desired. The following example (figure 12) is based
161

Figure 12.—Analysis of support system options when

convergence requirement is reduced from 4 to 2 in.

on a ground reaction curve selected from the program's

database. Using this curve, convergence is reduced from 4 in,

achieved with two rows of conventional 4-point, mixed

hardwood cribs on an 87-in spacing, to 2 in with alternative

supports.

1. Choose Ground Reaction Curve as the basis for selecting

the design criteria. Change the default entry height from 96 to

84 in. In the Ground Reaction form, the Emerald ground reaction curve is chosen. Input 4 in for the design convergence, and

the program will determine the required support load density

consistent with the ground reaction curve. In this case, it is

12.50 tons/ft.

2. Control will then be transferred to the Select Supports

window. Pressing the Add button brings up the Add Support to

Analysis window. Choose wood cribs as the support type.

3. Control will then be transferred to the Wood Crib

Performance window, where parameters for a 4-point wood

crib are defined (6×6×36-in timbers, mixed hardwood species,

two rows, two timbers per layer). The program will then

determine the installation requirements for this crib design as a

98.5-in spacing of the support. The program proceeds to the

Costs and Material Handling-Wood Cribs window. Review the

default settings for wood crib. The program then computes an

installed cost per foot of entry for this support design.

4. Control is then transferred to the Case Comparison

module. Review the performance parameters for this support

system.

5. The Comparison window is then closed by transferring

control back to the Main Menu. Activate Design Criteria and

edit the Ground Reaction Curve design criteria (Emerald Mine)

by changing the design convergence to 2 in. A warning message will be displayed, which indicates that the design criteria

have changed. The 4-point wood crib system must then be updated by pressing the Design and Cost button. The installation

requirements for this crib support system are changed to 44 in

for the new design criteria.

6. Control will then be transferred back to the Select Supports window. Select one of the various alternative support

technologies for analysis and enter the appropriate design parameters in the Performance window.

7. When the support designs are completed, the Case Comparison window will show baseline wood crib performance at

4 in of convergence and the alternative support performance at

2 in of convergence.

REPLACING TIMBER SUPPORTS WITH OTHER

ALTERNATIVES

There are several alternatives to timber supports. STOP can

be used to evaluate these various alternatives and make comparisons based on equivalent support capability or show the

advantages of alternative placement strategies with superior

roof support systems. The following example shows how these

alternative support technologies can be designed relative to the

current roof support system using available ground reaction

data.

The process begins by selecting the Current Support System

for design basis. The entry height is left at the default setting of

96 in. In this example, the current roof support system is a

double row of 4-point wood cribs. Thus, the user selects wood

cribs as the support type and enters the appropriate data in the

Wood Crib Specifications form to define a 4-point crib

constructed from 6×6×36-in timbers oak timbers. The centerto-center support spacing (108 inches in this case) and the

number of support rows (two in this case) are entered. Design

convergence and support load density are determined by

clicking on the Set Using GR Curve, where it is shown that the

current support system intersects the chosen ground reaction

curve at 3.34 in of displacement and provides a support load

density of 15.25 tons/ft. Table 1 compares several alternative

support systems that provide equivalent or improved support

capability. It is noted that with some supports, the support may

shed load prior to the design convergence. The program logic

is set to use the design convergence, but the user can determine

from the ground behavior and support performance curves the

convergence at which these supports will provide the required

load density (see example in table 1).

USING OPTIMIZATION ROUTINES TO SELECT

BEST SUPPORT DESIGN

The previous examples have shown how STOP determines

the required support spacing needed for a user-specified support

design. The optimization routines allow the user to specify support spacing and number of rows of support elements, and the

program will determine which support design best fits the load

and convergence design criteria. In the example shown in

table 2, the design criteria of 16.67 tons/ft of entry at 3 in of
162

Figure 13.—Comparison of conventional wood supports with three alternative support systems. A, Number of supports per supply

car; B, number of supports constructed per shift; and C, amount of construction work in foot-pounds.

Table 1.—Comparison of alternative support technologies as replacements

for conventional wood cribbing

Support type Design specifics

Installation

requirements Achieved

convergence

control, in

Achieved

support load

density,

tons/ft

No. of

rows Spacing

Wood cribs . . . . . . 4-point (6×6×36-in oak timbers) . . . . 2 108 3.34 15.2

Pumpable crib . . . . 30 in . . . . . . . . . . . . . . . . . . . . . . . . . 1 73 3.34 1

(0.24) 15.2

ACS . . . . . . . . . . . Pizza headplate . . . . . . . . . . . . . . . . . 2 92 3.34 1

(1.51) 15.2

Can . . . . . . . . . . . . 24-in-diam . . . . . . . . . . . . . . . . . . . . . 2 138 3.34 15.2

Stretch prop . . . . . Timber ft/hd boards . . . . . . . . . . . . . . 3 61 3.34 15.2

NOTE: Design requirements: support load density ' 15.25 tons/ft; convergence ' 3.4 in. 1

The required support load density of 15.2 can also be achieved at less displacement since the support sheds

load prior to the design convergence of 3.34 in. Using the mouse coordinates on the ground behavior and

support performance curve in the appropriate support design window, the convergence that produces the

required 15.2 tons/ft of loading can be determined. 

Table 2.—Support systems determined by the optimization routine

for user-defined support installation parameters

Support type Optimized design

User-specified installation

requirements Achieved

convergence

control, in

Achieved

support load

density,

tons/ft

No. of

rows Spacing

Wood cribs . . 9-point (5×5×30-in timbers) . . . . 2 96 2.8 17.6

Link-N-Lock 36 in . . . . . . . . . . . . . . . . . . . . . 1 96 2.6 18.5

Tri-Log cribs 30-in standard . . . . . . . . . . . . . . 2 96 1.5 23.0

Propsetter . . . 10-in-diam . . . . . . . . . . . . . . . . . 2 96 2.6 18.4

Can . . . . . . . . 24 in . . . . . . . . . . . . . . . . . . . . . 2 96 1.6 22.5

NOTE: Design requirements: support load density ' 16.67 tons/ft; convergence ' 3.0 in.

convergence were established from a ground reaction curve

chosen from the program database. A 96-in spacing was chosen

for the analysis by entering 96 inches in the Support Layout

section for the user-defined spacing option. Then various support

types can be selected for evaluation. When the Optimization

button in the Performance form is selected, the user will define

the installation requirements (number of rows and support

spacing), and the program will determine the support model that

most closely matches the design criteria (16.67 tons/ft at 3 in of

convergence). Since the installation spacing and number of rows

are specified, the achieved convergence will vary depending on

the support type chosen. Table 2 documents some examples of

optimized supports as determined by STOP.

EVALUATING MATERIAL HANDLING ASPECTS OF

SUPPORT DESIGN

Surveillance data show that material handling injuries are

common in support construction, resulting in several thousand

lost workdays each year. Thus, part of the support selection

process should be material handling requirements. Figure 13 is

an example of data derived from STOP for four support systems: (1) pumpable crib, (2) Alternative Crib Support (ACS),

(3) Propsetter, and (4) conventional 4-point crib. As seen from

this analysis (figure 13), there are significant material handling

advantages in using the alternative support technologies instead

of conventional wood cribbing.
163

CONCLUSIONS

Historically, wood cribs have been used as secondary roof

support in underground mines. The Wood Crib Performance

Model was developed in 1994 to provide an engineering

foundation for the design and applications of these supports. In

the past few years, many new support technologies have been

developed by various roof support manufacturers. In many cases,

these innovative support technologies provide superior roof

support and reduced material handling requirements. However,

each has its own performance characteristics; thus, they all need

to be employed differently to provide equivalent roof support.

STOP was developed to provide a more comprehensive support design program than that provided by the original Wood

Crib Performance Model. Not only does STOP include new

support technologies, it also allows for application of a new

design methodology based on a measured ground reaction curve

at a particular mine site. Also included in STOP are a comprehensive material handling assessment and cost evaluation for

each support system. STOP is a Windows-based program that

is user-friendly and very flexible and provides engineering

solutions for various secondary support applications.

STOP can be used to determine installation requirements for

an alternative support technology that will provide equivalent

support compared to a mine's current support system or an

installation that will provide a specified support load density at

a designated roof convergence. The optimization routines in

the program will select the most efficient support design for the

user-specified criteria.

STOP uses performance data developed by NIOSH through

safety performance testing in the Mine Roof Simulator. Each

support system has been evaluated through a rigorous testing

protocol that simulates in-mine service conditions. Photographs

of support conditions at various stages of loading are also

included in STOP. These photos help the user gain an understanding of the limitations of the support and can be used to

assess general loading conditions when these profiles are observed underground.

STOP can provide some much-needed engineering for secondary roof supports. This can be very helpful when petitioning MSHA for approval to use an alternative support technology or changing applications, such as increasing support

spacing. By proper engineering of the support relative to

ground reaction, convergence can be controlled to a predetermined level. This will allow an operator to optimize the

support application and provide a margin of safety in roof stability that will reduce the likelihood of roof failure without the

need for excessive roof support. Likewise, proper engineering

will remove uncertainty in support design and prevent the

application of inadequate support that can lead to roof falls.

Finally, STOP will allow mine operators to consider fully the

material handling aspects of support design in the selection

process, thereby reducing the incidence and severity of material

handling injuries.

Copies of the STOP software program can be obtained from

Thomas M. Barczak, NIOSH Pittsburgh Research Laboratory,

P.O. Box 18070, Pittsburgh, PA 15236-0070; phone: (412)

386-6557; e-mail: TBarczak@cdc.gov. It is also anticipated

that STOP will eventually be available through the NIOSH

Web site (www.cdc.gov/niosh).

ACKNOWLEDGMENT

The author acknowledges Dave Conover, NSA Engineering,

Inc., who developed the Windows-based program for STOP.

The STOP architecture evolved over a year of effort through

several painstaking revisions to form the final product. It is to

Mr. Conover's credit that this program was completed within

the budget constraints of the project and with such a high

quality of workmanship.

REFERENCES

Barczak TM, Gearhart DF [1993]. Engineering methods for the design and

employment of wood cribs. Pittsburgh, PA: U.S. Department of the Interior,

Bureau of Mines, IC 9361.

Barczak TM, Mucho TP, Dolinar DR, Bryja J, Bower J [1999]. Longwall

tailgate support, consideration, design, and experience. In: Proceedings of

Longwall U.S.A. Pittsburgh, PA, pp. 79-105.

Mark C [1992]. Analysis of longwall pillar stability (ALPS): an update.

In: Proceedings of the Workshop on Coal Pillar Mechanics and Design.

Pittsburgh, PA: U.S. Department of the Interior, Bureau of Mines, IC 9315,

pp. 238-249.

Mark C, Chase FE [1997]. Analysis of retreat mining pillar stability

(ARMPS). In: ProceedingsCNew Technology for Ground Control in Retreat

Mining. Pittsburgh, PA: U.S. Department of Health and Human Services,

Public Health Service, Centers for Disease Control and Prevention, National

Institute for Occupational Safety and Health, DHHS (NIOSH) Publication

No. 97-122, IC 9446, pp. 17-34.

Mucho TP, Barczak TM, Dolinar DR, Bower J, Bryja JJ [1999]. Design

methodology for standing secondary roof support in longwall tailgates. In:

Peng SS, Mark C, eds. Proceedings of the 18th International Conference on

Ground Control in Mining. Morgantown, WV: West Virginia University,

pp. 136-148.

Unal E [1986]. Empirical approach to calculate rock loads in coal mine

roadways. In: Peng SS, ed. Proceedings of the Fifth International Conference

on Ground Control in Mining. Morgantown, WV: West Virginia University,

pp. 234-241.
165

CABLE SUPPORT IN LONGWALL GATE ROADS

By Dennis R. Dolinar1 and Lewis A. Martin2

ABSTRACT

Cable bolt technology used by the U.S. coal industry was developed to a large extent in the 1990s. Today,

these cable systems include both cable bolts and cable trusses to provide supplemental and secondary support

in gate roads. This cable technology is significantly different than the cable systems in use in either U.S. hardrock mines or Australian coal mines. Development of this technology was initiated and spurred by research

efforts of the U.S. Bureau of Mines, and has continued under the health and safety programs of the National

Institute for Occupational Safety and Health (NIOSH). It was also followed up by work of roof support

manufacturers to create an essentially new support system. In this paper, the important support characteristics

of both cable bolts and cable trusses are discussed. The design of cable systems and the basis for that design

for tailgate and headgate situations are reviewed and explained. Case histories are presented on the application

of these support systems based on experience gained at a number of in situ test sites. 

1

Mining engineer, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA. 

2

Mechanical engineer, Spokane Research Laboratory, National Institute for Occupational Safety and Health, Spokane, WA. 
166

INTRODUCTION

Cable bolting was introduced into the U.S. coal industry in

1992 through the research efforts of the U.S. Bureau of Mines.

This original work was conducted at a longwall operation in

western Colorado for the purpose of finding alternative secondary

tailgate support [Tadolini and Koch 1993]. For the initial test,

cable design and installation were based on cable technology

developed for the hard-rock mine industry [Goris et al. 1993;

Goris 1990; Goris 1991; Goris et al. 1994]. This involved the

installation of the cables by hand and using a fully grouted cable

with a pumpable grout system. This technology was certainly

adequate for the hard-rock industry, but not for a high-production

longwall. Initially, a crew could install 8 to 12 cables per shift,

but with experience this number increased to 30 cables per shift.

From this beginning, the cable bolt system used today

quickly evolved until an essentially new product was created.

The result was a headed cable bolt installed by machine and

anchored with a resin grout cartridge and a partial grout column

[McDonnell et al. 1995]. Furthermore, resin manufacturers

developed special resins for cable bolt anchors. With these new

cable bolt systems, it is now possible to install up to 70 or more

cables per shift with a double-boom bolter. 

A driving force behind the development and use of cable

bolts as secondary support in the tailgate was to replace wood

cribs. Wood cribs, especially in the West, were becoming

expensive because of a shortage of timber and a lessening of

timber quality. A typical longwall gate road in the West

supported by wood cribs requires about 248 acres of timber

[Tadolini and Koch 1993]. Cutting this amount of timber has

an environmental impact. The installation of wood cribs is

labor intensive and materials handling is a significant problem,

not only from a logistic consideration, but also from an injury

standpoint. Especially in the West, where seam are high, a

large number of injuries result from crib installation. The high

density of the wood cribs necessary in a gate road restricts

ventilation and impedes the use of the tailgate as an escapeway

[Kadnuck 1994]. All of these considerations can put constraints

on the development of super longwall systems.

From a ground control standpoint, wood cribs are far from

ideal. Four-point wood cribs are regarded as a soft support

system. Improperly built cribs can result in a wide variation in

the performance of individual cribs, thus exacerbating ground

conditions. In many western mines, yield pillars are used, and

by design the tailgate will be subjected to large amounts of

deformation, much of which is the result of roof-to-floor

convergence. Yet crib systems will resist this deformation,

taking up a large portion of crib capacity. Wood cribs also have

problems handling the large lateral movements associated with

yield pillar designs.

Today, cable bolts are competing against other newly

developed types of standing support developed to replace the

wood cribs [Barczak et al. 1996; Mucho et al. 1999]. These

new systems are definite improvements over wood cribs. Still,

from a ground control standpoint these standing support

systems use a significant portion of their capacity to resist main

roof-to-floor convergence and limit full access by equipment to

the face. 

Cable bolts are used not only as tailgate support but also for

support in bleeders [Tadolini et al. 1993; McDonnell et al.

1995]. They are used as supplemental support, not only in

longwall operations, but also in room-and-pillar mines. There

are also efforts to adapt cable bolts as primary support because

they can be installed so they are much longer than the height of

the opening, a factor that could improve roof support in thinseam mines. Some consideration is also being given to using

a single-pass system in gate roads, where the cable bolts will be

used both as the primary support for development and secondary support when the panel is mined.

The Australians had used cable bolts in their coal mines as

both supplemental and secondary support [Gale 1987; Gale et

al. 1987]. The cables were normally 10 m (33 ft) long, fully

grouted, and installed with a pumpable grout system. Because

the Australians used a two-entry system with large abutment

pillars and 60 to 120 m (200 to 400 ft ) between entries, the

design and use of their cable system offered little that could be

adapted to use in a U.S. longwall tailgate. 

The development of cable bolts also spurred the development and use of cable trusses. Cable slings had been used

on a limited basis since the 1970s [Mangelsdorf 1982; Scott

1989]. However, anchoring the cables was a problem until a

system was devised that used anchorage systems based on a

resin cartridge inserted into the drill hole. These truss systems

can be installed with the assistance of a roof bolting machine,

though installation can still be accomplished in the traditional

manner with small drills. Cable trusses are now used as supplemental support in headgate entries, especially where horizontal stress causes damage. Cable trusses have been used in

open entry recovery rooms and to a limited extent, as secondary

support in tailgates.

This paper will discuss cable bolts and trusses and the design

of cable systems for the support of longwall gate road entries.

CABLE BOLT

Cable bolts are made from a high-strength steel cable. The

most common cable used is seven strands 1.52 to 1.59 cm (0.6

to 0.625 in) in diameter (Goris et al. 1994; McDonnell et al.

1995). The cable consists of six outer strands wrapped around

a middle or king wire strand (figure 1). The cross-sectional

area of the steel for the cable is 0.55 cm2

 (0.217 in2). Cable

bolts can be of any length, but typically range from 2.4 to 6.1 m

(8 to 20 ft) for use in a coal mine. The cables bolts are
167

Figure 1.–Seven strand steel cable used to make cable bolts.

Figure 2.–Cable bolt and cable bolt components.

anchored in the roof with resin grout cartridges using only a

partial grout column. This leaves a free cable length in the

lower portion of the hole. Cable diameters range from 1.27 to

2.29 cm (0.5 to 0.9 in), but only the properties of the 1.52-cm

(0.6-in) in diameter cable bolts will be discussed in the

following sections.

Figure 2 shows the components of a typical cable bolt. A

cable bolt consists of a cable head that ties the cable strands

together and allows the bolt to be installed and rotated with a

roof bolter. For ground control, the head is necessary for the

ungrouted portion of the cable to take load and resist rock

movement. The head also permits the installation of bearing

plates and other surface control devices. A barrel-and-wedge

system is used to attach the head on the cable. 

A stiffener is necessary to install the cable bolt and insert it

through the resin cartridge with a roof bolter. Without the

stiffener, the cable is too flexible to be pushed through the resin

cartridge and will bend outside the hole. If possible, the

stiffener should be long enough to be in the hole before the

cartridge is punctured by the cable and yet short enough to be

installed at a given mining height. A stiffener that is as long as

the resin cartridge will allow this to occur. Another function of

the stiffener is to prevent the cable from being nicked by the

bearing plate during installation, which reduces the potential for

corrosion of the cable. 

To assist in anchoring the cables and mixing the resin,

anchor buttons, "birdcages," nut cases, or bulbs are used in the

upper or anchor portion of the cable. These systems are designed for specific hole diameters, usually for holes 2.5 to 3.5

cm (1 to 1-3/8-in) in diameter. An end button holds the cable

end together and assists in inserting the cable through the resin.

A resin keeper or dam keeps the resin confined along the section of cable to be anchored and also compresses it. The necessity of the resin keeper will depend on resin viscosity, hole

diameter, and cable bolt design. 

Other designs allow the cables to be tensioned. This is

usually done at the head of the bolt through the wedge-andbarrel-head or by a threaded bar attached to the end of the cable.

Tensioning the head and wedge is done by hand while the

threaded rebar can be tensioned by the roof bolter. Cables with

yieldable heads are available where large roof deformation is

expected, and loads will exceed cable strength [Tadolini and

McDonnell 1998; Vandekraats et al. 1996].
168

Figure 3.–Cable bolt system installed in a tailgate.

Figure 4.–Load-deformation curve from a laboratory

tensile test of a cable.

Figure 5.–Load-deformation curve from a pull test of a cable

bolt underground.

CABLE SYSTEM CHARACTERISTICS

The cable system consists of a cable bolt with a ungrouted

or free length of cable and a resin grout anchor. The bearing

plate and other surface control devices held in place by the

cable bolt are also part of the system. The performance of the

cable support will depend on how well these components act

together as a system. Figure 3 shows a cable system installed

in a tailgate. In general, the cable should be the weakest part of

the system where the other components should be designed to

reach the ultimate strength of the cable. This includes the head,

anchor, and bearing plate. An exception is the specially

designed cable head that allow for controlled yield below cable

capacity.

CABLE BOLT CAPACITY

Cable ultimate strength will usually be between 244.7 and

266.6 kN (55,000 and 60,000 lbf) and will normally exceed 260

kN (58,600 lbf), while elongation of the cable at failure can range

from 3.5% to 8%. Cables will begin to yield at about 1% strain.

Figure 4 shows a typical test for a cable conducted in the

laboratory where ultimate strength exceeded 260 kN (58,600 lbf).

The load deformation curve from an underground pull test of a

resin-anchored cable bolt is shown in figure 5. In this pull test,

the cable length was 0.3 m (10 ft) and the grout anchor 0.9 m (3

ft). A maximum load of 268.7 kN (30.2 tons) was achieved with

the cable failing during the test, indicating that the anchorage

exceeded cable strength. In this case, two strands were broken on

the cable, which is typical failure for a cable. When a cable

breaks, there is a sudden, drop in load, often to near zero. This is

then followed by some load recovery, but this is limited by the

strength of the remaining strands and can be highly variable. The

load will drop again when the remaining intact strands begin to

break. Essentially, the final residual strength of the cable will be

zero although the cable will have some intermediate residual

strength. From pull tests, the elongation at failure is usually less

than 4% strain [Barczak et al. 1996].

CABLE SYSTEM STIFFNESS

The stiffness of a cable bolt will be determined by the free

cable length in the hole and the elongation properties of the

cable. However, elongation in the resin-anchored section of the

cable will influence stiffness and must also be considered. The

deformation properties of a cable consist of three components-a

construction, elastic, and rotational elongation. Construction

stiffness is permanent but is usually small. The rotational component is due to the rotation of the cable about the axis during

a test or as the cable is loaded. The elastic component is

dependent in part on the elastic modulus of the steel composing

the cable. The elastic modulus of the steel is 203.4 GPa (29.5

million lbf/in2

). However, the elastic modulus of the cable is
169

also dependent on the construction of the cable, which involves

the lay length. The lay length is the distance one strand takes

to make a complete revolution around the cable [Maryland and

American Iron 1985]. The stiffness can be calculated from the

following equations.

K ' E x A /(L) (1)

where K ' stiffness, kN/cm (lbf /in),

A ' area of cable, cm2

 (in 2 ),

L ' free cable length, cm (in), 

and E ' elastic modulus of the cable, GPa (lbf/in2

 ).

Knowing the elastic modulus, length, and area, the cable

stiffness can be calculated for a given load. Cable stiffness has

been measured underground using a pull test on cables installed

in a limestone roof [Zelanko et al. 1995]. For a 3-m-(10-ftlong) cable bolt with a 1.5-m (5-ft) resin anchor and therefore

a 1.5-m (5-ft) free length of cable, the initial cable stiffness

below the system yield was 106 kN/cm (30.4 tons/in) for cables

installed in a 2.5-cm (1-in) in diameter hole, and 98 kN/cm (28

tons/in) for a cable installed in 3.5-cm (1-3/8-in) in diameter

hole. 

Based on these stiffness values, the deformation modulus

of the cables can be calculated from the stiffness equations.

However, a correction must be applied to the free length of

cable to allow for elongation of the cable in the anchor. From

the load transfer characteristics and distances determined

experimentally for grouted rebar, the elongation of the cable in

the anchor can be approximated by an additional 20 cm (8 in)

of free cable length [Serbousek et al. 1987]. Although the anchors will affect the load transfer, any error in determining the

additional free cable length from cable stretch in the anchor

portion of the cable will have only a small effect on stiffness

calculations. From the above test results, for the 106-kN/cm

(30.4-tons/in) stiffness, a free cable length 1.72 m (5 ft 8 in) ,

and an area of 0.55 cm2

 (0.217 in2), the calculated cable

modulus is 132 GPa (19.1 million lbf/in2

).

Using this calculated elastic modulus and the stiffness

equation, the stiffness of cables bolts with different free lengths

can be determined. For a 4.3-m (14-ft) cable with a 1.2-m (4-ft)

anchor and 3.0 m (10 ft) of free cable length, cable stiffness

would be 56.4 kN/cm (16.1 ton/in). The assumption is made

that the anchor has sufficient length where the anchor will not

slip and a portion of the anchor will have little or no load below

the yield of the system.

The stiffness of the support will determine how quickly the

support will develop resistance and load as the roof deforms.

The cable bolt stiffness can be compared to the stiffness of

other support systems. For a 1.5-m (5-ft), long No. 6, fully

grouted rebar bolt, the stiffness is 700 kN/cm (200 tons/in), and

for a 1.8-m (6-ft) long, 1.9-cm (3/4-in) in diameter point-anchor

system , the stiffness is 175 kN/cm (50 tons/in) [Karabin and

Hoch 1979]. 

Cable bolts have much less stiffness than most primary support systems. Although the cables are more flexible, the lower

stiffness indicates that they will not resist movement as much as

other primary support for a given load. For secondary support, a

four-point poplar wood crib is 1.8 m (6 ft) high will have a

stiffness of 75.3 kN/cm (21.5 tons/in) [Mucho et al. 1999]. This

is equivalent to a cable bolt with a 2.3 m (7.5 ft) free length.

However, in a tailgate support system, at least two or three cable

bolts would be used in place of a single crib. In this case, the

cable system would be two or three times stiffer than a crib.

SHEAR CHARACTERISTICS 

Resistance to shear and lateral movement can be developed

along the free length of the cable and result in cable loading.

To determine the shear characteristics of the cables, a series of

laboratory tests were conducted where both ungrouted and

grouted cables were installed across the block boundary in pairs

of concrete blocks [Goris et al. 1995; Goris et al. 1996]. The

blocks were sheared parallel to the contact surface and perpendicular to the installed cable bolt. The results showed that

the initial peak shear strength was not changed, but that the

residual shear strength at 3.8 cm (1.5 in) of displacement was

doubled (figure 6). The cables were not immediately activated,

but required about 1.0 cm (0.4 in) of displacement before

resisting the shear for a 3.5-cm (1-3/8 in) in diameter hole and

about 1.52 cm (0.6 in) before significant resistance occurred.

Essentially, shear is resisted only when sufficient movement

has occurred and the roof has already been mobilized.

In this series of tests, the maximum lateral displacement on

the cables was about 3.8 cm (1.5 in). None of the cables failed

as a result of this level of displacement. At this point, the

cables were loaded to about 60 kN (13,500 lbf), still well below

the ultimate strength. However, in a field study, where about

0.3 to 0.46 m (1 to 1.5 ft) of lateral movement occurred across

the entry, several cable bolts had failed. It is estimated that the

cable bolts failed at between 5 to 10 cm (2 to 4 in) of lateral

movement [Dolinar et al. 1996]. The failures occurred from a

combination of shear and tension. 

RESIN ANCHOR 

Several factors influence the effectiveness of the resin

anchor, including anchor length, type of resin, and hole

diameter [Zelanko et al. 1995]. Figure 7 shows a cross section

of a cable installed in resin. The cables will transfer the load

through the anchor to the rock with all the load transfer taking

place within the anchor length. However, if the anchor is too

short, the anchor could slip (rock-grout interface failure),

especially in weak rock. With a longer anchor, the anchor
170

Figure 6.–Typical cable load and shear stress versus shear displacement A, smooth joint with cable

bolt reinforcement and B, rough joint with cable bolt reinforcement.

Figure 7.–Cross section of 1.52 cm (0.6 in) cable set in resin,

installed in a 2.5 cm (1 in) diameter hole.

could still fail along the grout-cable interface as the cable is

loaded and yields where the anchorage will not exceed the

strength of the cable. The anchor failure mechanisms are

discussed in more detail elsewhere [Goris 1990; Goris 1991].

An important aspect to the development of an adequate

anchorage is the addition of buttons, nut cases, garford bulbs,

or birdcages to the anchor portion of the cable. During cable installation, these anchor components assist in mixing the resin,

which should provide for an improved quality and consistency

of the resin anchor, especially in the larger-diameter holes (3.5

cm [1-3/8 in]). Further, laboratory pull tests on short (76.2 cm

[30 in and less) column grout anchors have shown that anchor

components embedded in the grout significantly increase

anchorage capacity over that of a cable without embedded

anchors [Goris 1990, 1991]. With an increase in the resin

column length to 0.9 m (3 ft), the conventional cable without

anchors can achieve the ultimate strength of the cable, although

test results are somewhat inconsistent. In laboratory pull tests,

only 60% of the tested cables reached the ultimate cable

strength while 40% of the cable anchors failed at significantly

lower loads [Martin et al. 1996a]. Without the addition of anchor components embedded in the resin, there is a high

probability that the cable anchor will be significantly weaker

than the cable.

Both laboratory and field investigations using pull tests

have shown that 1.2 and 1.5 m ( 4 and 5 ft) of resin anchor will

achieve the ultimate capacity of the cable if properly grouted

[Martin et al. 1996a; Zelanko et al. 1995]. Although laboratory

tests have shown that a length of 0.9 m (3 ft) or less of anchor

can result in the cable reaching ultimate strength, this was not

achieved on a consistent basis. In the elastic range, below the

yield of the cable, most of the load is transferred within the first

0.6 m (2 ft) of the anchor [Goris 1990; Goris 1991; Serbousek

et al. 1987]. Essentially, the cable load in the anchor decays

exponentially with distance along the anchor. Beyond about

0.6 m (2 ft) there will be little load in the anchor. However,

once the cable yields, the lower portion of the resin anchor may

begin to fail or the cable may debond from the grout, resulting

in the loads being transferred further up the anchor. Essentially,

the anchor becomes shorter as the cable is loaded beyond yield.

Therefore, a minimum of 1.2 or 1.5 m (4 or 5 ft) of resin will

provide a margin of extra length in allowing the cable to reach

the ultimate strength. Also, this margin of extra length gives

some degree of safety for improper grout installation although
171

Figure 8.–Surface control in the form of a "Monster Mat" installed with cable bolts.

this is no guarantee that even a 1.2- or 1.5-m (4- or-5 ft) long

anchor not properly installed will result in the cable reaching

ultimate strength. Cable systems using longer resin anchor columns of 2.1 to 2.7 m (7 to 9 ft) have also been installed.

However, depending on the resin used and the thrust capacity

of the bolter, problems may be encountered with inserting the

cable through that length of resin column. 

An investigation was conducted underground using pull

tests on 3-m (10-ft) cable bolts with 1.5-m (5-ft) resin anchors

to evaluate parameters other than length that could affect the

anchor performance [Zelanko et al. 1995]. Parameters varied

in this study included hole diameter, resin type, and use of a

resin keeper. The 0.6-in cable is generally installed in either a

2.5 or 3.5-cm (1-or 1-3/8-in) in diameter hole. Overall, the

capacity and stiffness of the installed cable bolts were lower in

the larger-diameter holes (3.5 cm [1-3/8 in] as compared to a

2.5 cm [1 in]. Although an adequate anchor can be achieved in

holes of either diameter, there is less consistency in performance with the larger holes.

Higher-viscosity resins gave a better anchor performance

than lower-viscosity resins. Cable resins are less viscous than

the standard resins and allow for easier installation with the roof

bolter. However, besides decreased performance, resin loss

could occur more easily with these types of resin, thus requiring

longer column lengths. More viscous bolt resins can be used,

but with longer resin lengths, older bolting machines may not

be able to thrust the cable through the cartridge. Furthermore,

with the standard resins, another installation problem that can

develop is loosening of bearing plates or plates not being in

contact with the roof. In such cases, the higher- viscosity resins

prevent the end of the cable from being pushed up completely

through the cartridge even though the roof bolter was able the

push the plate against the roof. When this occurs, the ungrouted

section of cable bends and with the release of pressure, the

cable will spring back, resulting in a loose plate. Resin keepers

were also found to be important for the larger-diameter holes,

but not a factor in 2.54-cm (1-in) in diameter holes. However,

the need for a resin keeper will depend on the overall design of

the anchor section of the cable bolt.

From this study, the system with the thinner annulus, the

1.52-cm (0.6-in) in diameter cable in a 2.5-cm (1-in) hole, performed better with more consistent behavior than the system with

the larger annulus. In this regard, the button, birdcage, or nut case

diameters must be matched with the proper hole size. This study

highlights not only the necessity for a properly installed cable

anchor, but also a properly designed anchor system as well. 

CORROSION

Corrosion is a issue, although the extent of the problem is

not completely known. Cables are more susceptible to corrosion and failure than other types of support. A nick in a cable

strand that corrodes has a much greater impact than corrosion

in a roof bolt. Some observations suggest that bright or black

cables have about a 10% decrease in area of the strands six

months after installation [Martin et al. 1996b]. Both galvanized

strand and epoxy coated cable can be used to minimize the

potential for corrosion [Goris et al. 1994]. Manufacturing techniques are now available that do not adversely effect either the

strength or flexibility of galvanized cable [Tadolini et al. 1994].
172

SURFACE CONTROL

Bearing plates are necessary for the functioning of a cable

bolt with a free length and allow the cable to load and resist

rock movement while transferring this load thorough the anchor

to rock deeper into the roof. Therefore, the bearing plate must

be designed for the cable to reach ultimate strength, or a

minimum of 260 kN (58,600 lbf). "Monster Mats" and T 5

channel are often installed with the cables to provide additional

support and surface control across the row of cables (figure 8).

A Monster Mat is a steel pan 0.48 cm (3/16 in) thick and 33 cm

(13 in) wide, while a T 5 channel is 0.5 cm (0.2 in) thick and

has a 10-cm (4- in) wide bearing surface. Both systems can add

significantly to surface control and also provide some structural

support. These systems are installed in conjunction with the

high-capacity bearing plates.

DESIGN CONCEPTS FOR TAILGATE SUPPORT

In the tailgate, the primary support is designed to withstand

development mining, but may not be able withstand the

longwall environment and control the lower roof. Therefore,

cable bolts can be installed as secondary support to maintain the

entry. The cable bolts must keep the roof from falling and the

entry open during panel mining. As the lower roof moves and

deforms, the cables will distribute the forces that develop below

a given failure horizon deeper into the roof through the cable

and anchor support. Although there is primary support, it is not

normally taken into account when designing the cable support

system or, for that matter, another type of secondary support

system. 

The basic design concept in using partially grouted cable

bolts to support the roof is suspension. Essentially, the cable

bolt system must maintain and control the dead weight load of

rock or rock movement below a potential failure horizon in the

mine roof. This in part determines the spacing of the cable

bolts. Furthermore, an adequate cable anchorage length must

be obtained above a given failure horizon and, combined with

the location of the failure horizon, determines cable length.

Experience based on test sites in tailgates have further refined

and established a basic design for cable spacing and row spacing. Although the cable systems are designed for the full dead

weight of the rock, this is seldom seen and is somewhat an

oversimplification of conditions, but it provides a starting point

for design and designing to a worst-case scenario. Also, lateral

roof movement, as well as vertical expansion from lateral roof

movement, can cause significant loads to develop on the cables

even beyond the weight of the rock.

CABLE LENGTH

The selection of cable length is the probably most crucial

aspect of the design of a cable system. Depending on geologic

conditions, selecting a length may be simple and straightforward, while in other cases, it may require an iterative process

using a range of information. The key is to identify the location

of potential failure horizons in the roof that may develop when

the panel is mined. 

Once the deepest potential failure horizon is identified, the

cable length will be the depth of this failure horizon plus the

length of an adequate anchor. Typical cable lengths in gate

roads are between 3.7 and 4.9 m (12 and 16 ft). However, a

minimum length in general should be for the cable bolt to be

long enough to be anchored above the primary support. In this

case the primary support zone is being suspended by the cables.

However, there may be failure planes that develop above the

primary support and require a longer length of cable bolt. This

potential failure zone may be a flat or arched surface, depending

on how the roof may fail. In a gate road situation, much deeper

movements may occur that are not relevant to the stability of

the immediate roof or the opening. 

The initial step in designing an adequate support system

requires gathering detailed information on ground conditions

and the underground mining environment. To determine a

potential failure horizon will require examining the roof and

roof geology or evaluating roof performance to determine an

adequate cable length. Such information may include a general

estimate of rock mass strength or rating, geologic structure, and

strengths of the immediate and main roof members. This

information can be obtained from roof core samples and

supplemented by observations from a borescope or camera to

evaluate test holes in the roof. If the rock overlying the

immediate roof is stronger or more competent, this may be an

obvious place to locate the anchorage and is the easiest situation

for determining cable length. However the geology may not be

that clear-cut or the depth of the stronger unit may be too deep

to be of practical use for supporting the immediate roof. Actual

mining experience, test sites, and examination of roof falls can

provide more data to help in the design of the cable system.

Tests sites with instruments such as multi point extensometers

can also be used to locate and evaluate these potential failure

horizons. Such instrumented test sites can be used to confirm

the adequacy of cable's length and design. 

DESIGN FOR SUSPENSION

For cables, to consider that the rock is being supported

through suspension may be an oversimplification, but does

provide a basis for establishing the initial design of the system.

Designing for suspension requires that the cables carry the

weight of the rock under the potential failure zone, which, in
173

Figure 9.–Detached block of failed roof supported by cables.

many situations, is the worst case scenario [McDonnell et al.

1995]. In some situations, there will be loads that actually

exceed rock load because of geology, horizontal stress, lateral

rock movement, and mining-induced loads. 

For suspension, the simplest approach is to identify a

parting plane or a flat-lying, potential failure plane above the

bolted roof horizon where the roof will shear at the pillar edge

of the opening and the entire weight of the rock must be

supported as a detached block (figure 9). The weight of the

material can be determined from the following equations.

Fw ' We Hp , (2) γ

where Fw ' weight of rock per linear length, kN/m (lbf/ft),

We ' effective width of opening, m (ft),

Hp ' distance from coal roof to parting plane, m (ft) ,

and ' rock density, kN/ m3

 (lb/ft3 γ ).

If an arched roof failure is formed with the pillars carrying

some of the weight, the cables need only support the weight of the

rock under the arch (figure 10). The height of the arch will be determined by a combination of the geology, as well as by the vertical and horizontal stresses acting on the roof and the induced

mining stresses. Obviously, the length and the number of cables

will depend on the height of the arch, and therefore this requires

the identification of the failure surface. The weight of the material within the arch can be estimated from the following equation.

Fa ' π

4

we Ha γ , (3)

where Fa ' weight of rock under pressure arch per linear

foot, kN/m (lbf/ft),

and Ha ' height of pressure arch, m (ft).

The behavior of the pillar under different loading conditions will affect the width of the opening and therefore the

weight of the rock that must be supported (figure 11). The

depth of the yield zone can be determined from equations

developed by Wilson and depend on the strength of the coal

pillar [Wilson 1972]. The following equations can be used to

estimate the depth of the yield zone. w ' pillar width in meters

(feet).

(1) Rigid floor conditionsB

w ' 2 m

F

ln q

p % pN (4)
174

Figure 10.–Formation of pressure arch of failed mine roof material.

Figure 11.–Formation of yield zone in coal pillar, (We ' effective width of opening; yp1 ' yield zone of pillar 1; yp2

' yield
175

w ' m q

p % pN

1

k & 1

&1 (5)

F ' k & 1

k

% (k & 1)2

k

tan&1 k , (6)

Figure 12. –Design chart to determine yield zone width in

coal pillars. A, yielding roof; B, rigid floor.

Figure 13.–Cable support design chart, where the effective

width of opening (We )' 7.6 m (25 ft) and rock density ( ) γ ' 2,403

kg/m3 150 lb/ft3 [McDonnell et al. 1995].

(2) Yielding roof-floor conditionsB

where tan&1

/k is expressed in radians,

m ' seam height, m (ft),

q ' overburden load, t/m2

 (st/ft2),

p ' artificial edge restraint, 0 t/m2

 (st/ft2),

pN ' uniaxial strength of fractured coal, 1/m2

 (st/ft2),

and k ' triaxial factor ' , where N ' angle of 1 % sin φ

1 & sin φ

interval friction, deg.

Figure 12 shows charts developed from these equations to

calculate the depth of the yield zone. The charts were created

using a angle of internal friction of 35E. The effective opening

or roof width can then be determined from the following

equation:

We ' W + Yp1 + Yp2, (7)

where W ' mined width of opening, m (ft),

Yp1 ' yield zone for pillar 1, m (ft),

and Yp2 ' yield zone for pillar 2, m (ft).

Based on the weight of material that must be supported, the

spacing of cable bolts across the opening, as well as row

spacing, can be calculated. Using a cable with a capacity of

260 kN (58,600 lbf) and varying the number of cables across

the opening and row spacing, a design can be determined for

different thicknesses of rock that must be supported. Figure 13

shows this design chart for an effective width of 7.6 m (25 ft)

and a rock density of 2,403 kg/m3

 (150 lb/ft3). This chart is

based on a flat failure surface developing at the given horizon

with the additional weight of material for the yield zone. A

separation at 2.4 m (8 ft) would require four cables per row

with 2.4-m (8-ft) row spacing. However, there are no safety

factors calculated into these charts.

PERFORMANCE OF CABLE BOLTS WITH RESPECT

TO TAILGATE INTERACTION ZONES

When using cable bolts for secondary support in the

tailgates, there are three zones that must be considered in

evaluating the design and performance of the cable systems.

These zones are the outby abutment zone for both vertical and
176

Figure 14.–Tailgate abutment zone outby the face supported with cable bolts.

horizontal stress, the shield zone from the face to the back of

the shield, and the cave zone. Each zone has different

performance requirements, and therefore, the cable system

much be designed to meet these requirements. If problems do

occur in the tailgate that results in the shutdown of the face, the

cost to the operation in both downtime and clean-up can be

high.

In the forward abutment zone, the cable support must maintain an open tailgate entry and prevent any major roof falls that

impede the use of the tailgate as a secondary escapeway and for

ventilation (figure 14). In the abutment zone, the cable loads

will depend in part on the geology, depth, and pillar design.

This zone receives the most support from the pillars and the

panel and may be up to 45 to 60 m (150 to 200 ft) wide. The

depth will control the pillar yield zone that develops along the

tailgate entry and at the face, where an increase in the entry or

intersection span will, in general, result in more roof separation

and movement. This yield zone will obviously increase near

the face. With pillar design, abutment pillars will offer the most

support to the tailgate entry. In many situations, little load or

roof movement will be seen. With a yield pillar adjacent to the

tailgate, significant roof movements and cable loads can develop when the pillar yields. Often this will include lateral

movement that the cables must withstand.

Geologic structures such as joints, faults, and sand channels, can cause locally high loads to develop on the cable support and can result in cable failures and even small roof falls.

These roof control problems will usually begin with a sudden

increase in the rate of vertical loading from the abutment. In

such cases, some additional support may be required locally if

the cables fail. Although horizontal stress is not typically a

problem in the tailgates because of the adjacent caved panels,

horizontal stress damage to the roof may have been caused by

a previously mined panel and the damage may have been

transmitted through the crosscuts to the tailgate entry. Lateral

roof movement may occur just outby the face and result in

additional cable loads or even cable failure in shear. Furthermore, damage done to the roof in this zone and subsequent

loads on the cables will impact performance in the other zones.

In general, in the abutment zone, the highest loads and roof

movements will be seen in the intersections although with yield

pillars, this may occur at mid-pillar. 

In the zone from the face to the back of the shields, performance requirements are very similar to those for the

abutment zone Cthe area must remain open as an escapeway

and for ventilation (figure 15). However, support of the panel

has been removed and replaced by the shields, and this creates

an opportunity for the roof to move because of the loss of

support. Therefore, higher cable loads will develop here than

in the abutment zone. This is the situation for which the cable

system should be designed. The degree of roof movement and

separation will depend to a large extent on the geology and any

previous damage done to the roof. Cables often begin to load

in this zone when there was little movement in the abutment

zone, especially when abutment pillars are adjacent to the tailgate. Maximum loading and roof movement are seen just as the

cables go behind the shields. 

In many operations, there is no need to maintain the tailgate

behind the shields and the performance of the cable bolts in this
177

Figure 15.–Tailgate shield zone supported by cable bolts.

area is not a factor. However, in some mines, ventilation requirements necessitate that the gateroad be kept at least partially

open to the nearest crosscut behind the face, a distance of

usually 30 to 45 m (100 to 150 ft). The maintenance of this

section of the tailgate by the cables is dependent to a large

degree on the geology and the cave and only to a limited degree

on cable system design. In the tailgate adjacent to the cave, the

roof develops into a cantilever that must be supported. If the

roof is not strong enough and the cave goes above the cables

then the cantilever could fail and close most of the entry (figure

16). If the roof is strong enough to maintain the cantilever, then

the cables will help to maintain any lower weaker roof. The

critical factors are whether the cave develops above the cables

and if the zone is strong enough to maintain the cantilever. The

cables probably add little overall strength to the cantilever.

However, there are cases where the entry has stayed open more

than 45 m (100 ft) behind the face [Koehler et al. 1996; Martin

et al. 1996; Mucho et al. 1996] (figure 17). 

Geologic structures such as joints can cause periodic failure

of even a competent roof behind the face. Essentially, there is

no guarantee with cables that this zone can be maintained to the

next crosscut. If the tailgate must be kept open, then other

types of support should be considered. However, even if the

roof fails, a portion of the tailgate alongside the pillar will

usually remain open, although this is a restricted area [Molinda

et al. 1997].

DESIGN BASED ON TEST SITES 

Test sites have been used to establish, evaluate, and

confirm cable system designs. Besides being a good practice,

test sites may be required by MSHA when cable systems are

used for the first time at a mine. Test sites can also be used to

modify existing cable system designs. Although observation

can be used to judge the successes of the design, instruments

that monitor both roof movement and separation and cable

loads to quantify the results and confirm the design are preferred. Monitoring of roof movement is especially useful when

evaluating cable length. Final cable system designs should be

based on evaluation of test sites. 

The design most used in tailgates has been one in which

there are four cable bolts per row. Although three bolts can

provide adequate support with the same safety factors, four

bolts per row have certain advantages. This number provides

good coverage across the entry, thus maintaining an effective

support front, especially as the cable row goes behind the face.

Also, in a given row, the failure of a single cable represents a

loss of support of only 25% with four cables per row and 33%

for three cables per row. Although the cable support is designed on the basis of an area of support, as the support goes

behind the face, the performance of a single row or the line of

support becomes important. Finally, with the use of doubleboom bolters, it is usually more efficient to install four bolts
178

Figure 16.–Tailgate behind the shields has caved though

supported with cable bolts.

Figure 17.–Tailgate behind the shields kept open with cable bolts.

than three bolts per row. Row spacing has varied from 1.2 to

1.8 m (4 to 6 ft). With row spacings wider than 1.8 m (6 ft),

interaction between rows can be lost and the effectiveness of

the reinforcement as a system reduced. 

Additional support to the crosscuts must also be considered

when using cable bolts because of the increased spans in the

intersections and any damage in the crosscuts from previous

panels. Generally, this support can consist of one or two rows

of cables installed in the crosscuts. Instead of (or in

conjunction with) the cables, cribs can also be set in the

crosscuts. Another modification to the design is to angle the

outside cable bolts toward the pillars and panel. This angle is

usually about 10E from vertical and will allow the anchorage to

be in a more stable roof zone.

DESIGNS FOR LATERAL MOVEMENT

Cable bolts will offer resistance to lateral movement, although shear is resisted to a large degree only after the peak rock

strength has been exceeded. Essentially, the rock has failed and

is now mobilized where the cables will offer significant postfailure resistance by significantly increasing the residual shear of

the rock [Goris et al. 1995, 1996]. However, in some cases, because of large lateral deformations, the cables may not be able to

stop or limit this displacement prior to failing. At a mine in

western Colorado, a tailgate supported with cable bolts was

subjected to large lateral deformation. This occurred as the adjacent panel was being mined, with the horizontal stress abutment

in the headgate causing roof damage not only to the headgate,

but also to the tailgate of the next panel through the crosscuts

[Dolinar et al. 1996]. This panel was supported with cable bolts

and rigid trusses. About 0.3 to 0.45 m (1 to 1.5 ft) of lateral

movement occurred in places along this entry. All the rigid truss

cross bars had been thrown from the anchor bolts while about

20% of the cable bolts failed. It is estimated that the cables

withstood about 5 to 10 cm (2 to 4 in) of lateral movement before

failure. These are very tough ground conditions where few support systems could be expected to prevent movement of this

magnitude. With shear or lateral movement, the flexibility of the

cable bolts is not fully utilized. 
179

Obviously, it may be difficult or impossible to stop such

large movements with support, and other approaches may need

to be considered to prevent support failure. If the support does

not fail, then it can support the damaged roof by suspension.

However, there are some alternative approaches that can be

used to minimize the impact of large lateral movements on

cable supports. One approach is to keep the cable bolts out of

the highest zones of shear or differential lateral movements that

occur near the edge of the pillar. To do this, cables can be positioned 0.6 to 0.9 m (2 to 3 ft) from the rib. Another successful

approach is to use cable bolts with a yielding head. These

heads will allow the cable system to yield in a controlled

manner at loads below the ultimate capacity of the cables

[Tadolini and McDonnell 1998; Vandekraats and Watson

1996]. Some of these heads will allow up to 50 cm (20 in) of

controlled movement, thus letting the cable deform with the

roof. With nonyielding cable heads, the head will lock in the

bolt, and stretch in the system must take place as the bolt goes

into a yield condition. 

CASE HISTORIES OF CABLE BOLTS AS SECONDARY SUPPORT IN TAILGATES

The following section gives case histories for tailgates

supported with cable bolts either as the main or only secondary

support system. In each of these cases, a 1.52-cm (0.6-in) in

diameter cable bolt with an ultimate capacity of 260 kN (58,600

lbf) was used. At these test sites, cable loads were monitored

usually with hydraulic U-cells and pressure pads, while differential roof sag measurements were made within and above

the cable horizon. Roof-to-floor convergence measurements

were also obtained at some sites. Usually, several intersections

as well as midpillar locations were monitored.

CASE HISTORY 1

This mine is located in western Colorado and used a yieldabutment pillar configuration. Three different cable bolt system

designs were tested in a 274-m-(900- ft-long) section of the

tailgate. They included a passive system, a stiff passive system

(increased grout anchorage length), and a tensionable system

[McDonnel et al. 1995; Tadolini and Koch 1993; Tadolini and

Koch 1994]. The roof geology consisted of 1.2 m (4 ft) of coal

overlain by 0.6 m (2 ft) of silty shale and 1.2 m (4 ft) of

interbedded shale, silty shale, and sandstone. After evaluating

the geologic data on the roof, it was thought that roof separation

would most likely occur in the silty shale although separation

might also develop higher in the interbedded shale and

sandstone. Above the immediate roof was a 4.9-m-(16- ft-)

thick massive sandstone. This sandstone provided a good anchorage from which to suspend the lower roof. The entry width

was 5.8 m (19 ft), but with pillar yield, the effective width for

design was assumed to be 7.9 m (26 ft). 

Figure 18 shows a tailgate entry cross section with the

cable configuration where four cables per row were installed on

1.5-m (5-ft) row spacings. The cable bolts were 4.9 m (16 ft)

long. With this configuration, the cables would have just

enough capacity to hold up 3 m (10 ft) of rock if the separation

occurred at this level and the full weight had to be supported by

the cables. For surface control, bearing plates, monster mats,

and wire mesh were used. For the passive site, the cables were

installed with a resin grout length of 1.7 m (5.7 ft), which

assured adequate anchorage in the sandstone and resulted in a

free cable length of 3.1 m (10.3 ft). For the stiff passive

system, the resin anchor length was 3.7 m (12 ft), leaving only

1.2 m (4 ft) of free cable length. In the tensional section, the

resin length was again 1.7 m (5.7 ft) with a free cable length of

3.1 m (10.3 ft). These cables were tensioned to 35 kN (8,000

lbf). Because of the thrust from the roof bolter, the cables in the

passive sections were installed with 6.7 to 22.2 kN (1,500 to

5,000 lbf) of load.

With panel mining in the passive area, the maximum total

roof separation was about 0.6 cm (0.25 in) in an intersection.

In the stiff and tensional areas, the maximum total separation

was between 3.2 and 3.8 cm (1.25 and 1.5 in) in both sections.

The movement and separation took place within 30 m (100 ft)

of the face and did not affect functioning of the tailgate or load

the support beyond the cable's strength. Cable loads in the

passive section ranged from 0 to 107 kN (0 to 24,000 lbf) and

averaged 21.3 kN (4,800 lbf). In the stiff section, cable loads

alongside the shields ranged from 71 to 116 kN (16,000 to

26,000 lbf). For the tensioned cable site, the loads ranged from

18.2 to 151 kN (4,100 to 34,000 lbf). However, in the

tensioned test site area, several geologic features, including coal

spars and a clay dike, were observed in the roof. In one small

area, the cables were loaded to over 133 kN (30,000 lbf), while

the roof was broken and fractured. In this area, some cables

appeared to have failed or the cable heads had slipped. Nine

wood posts were set to provide additional support, although the

section through the area was mined without incident. In the

passive area, the roof remained open 30 to 45 m (100 to 150 ft)

behind the face, while for the stiff system, the entry remained

open about 30 m (100 ft) behind the face.

All three systems worked extremely well and were able to

keep the tailgate open through the abutment zone, alongside the

shields, and even for a distance behind the shields. However,

from these test sites, it could not be determined if there were any

difference in performance among the systems. In the tensional

area, localized geologic structure in the immediate roof did induce

higher cable loads and possibly cable failures, but no significant

problems were apparent in controlling the roof. 
180

Figure 18.–Cross section of tailgate entry showing cable bolt design for case history 1.

CASE HISTORY 2 

Case 2 was a mine located in Utah with a double entry

yield pillar configuration in the tailgate [Tadolini and

Trackemas 1995]. The depth of cover at the mine averaged

about 460 m (1,500 ft). Because of the yield pillar, the effective

opening width was estimated at 9.7 m (32 ft). The geology of

the immediate roof consisted of thinly bedded siltstones, sandstone, and mudstones along with carbonaceous material to a

depth 1.2 to 1.8 m (4 to 6) ft. This was overlain by a sandstone

containing bands of carbonaceous material. Sand channels cut

into the immediate roof, but not into the coal, and affected roof

quality locally. 

Figure 19 shows the geology as well as a cross section of

the entry with the cable system design. The cable bolts were

4.9 m (16 ft) long with four bolts per row and a row spacing of

1.5 m (5 ft). The resin anchor length was 1.5 m (5 ft), resulting

in a free cable length of 3.4 m (11 ft). This free cable length allowed for greater cable elongation in the high-stress and deformation environment caused by crushing of the yield pillar.

This cable system design would support a roof thickness of up

to 3.3 m (10.9 ft) based on a dead weight load. 

The installed cable loads averaged 15.1 kN (3,400 lbf).

With panel mining, cable loads ranged from 0 to 178 kN (0 to

40,000 lbf) during the life of the test site. Loading and unloading of cables occurred in the same row, while shearing in

the roof was observed at different depths. This shearing action

resulted in differential lateral movement between roof layers

and could explain the loading and unloading of the cables. In

the area of the sand channels, several cables failed because of

this differential movement, and some standing support was

added. Separations were observed in the mine roof, but never

above the anchor horizon. Up to 10 cm (4 in) of overall roof

separation was seen, most of which occurred between the mudstone-sandstone layers within the lower 1.2 to 1.8 m (4 to 6 ft)

of the roof. This was within the elongation capacity of the

cables. However, cable loads up to 275 kN (40,000 lbf) and

cable failures indicated that this level of elongation was

approaching the limit of the cable system especially as it was

developed by shear. Despite this high-deformation environment, the tailgate was kept open and functional with the cable

support even under the sandstone channels (figure 20). 

In a series of initial experiments with cable support, the

mine installed a double row of wood cribs with spacing that

was increased from 1.8 to 6.1 m (6 ft to 20 ft) through the

tailgate test area. Finally, a section with no cribs and only cable

support was tested. The results of these trials indicated that the

best roof conditions were when there few or no cribs. The

hypothesis was that the standing support damaged the roof as

it resisted the main roof-to-floor convergence. The roof

damage and subsequent hazardous conditions resulted as the

cribs were compressed against the roof with such force that it

caused the immediate roof to break.

CASE HISTORY 3

Case 3 is a mine in western Colorado using a three-entry

system with two abutment pillars [Dolinar et al. 1996]. The

roof generally consists of a thinly bedded siltstone (stack rock)

and massive, fine-grained sideritic siltstones that grade laterally
181

Figure 19.–Cross section of tailgate entry showing cable bolt design for case history 2.

to a dark gray limestone and to sandstone. Another seam overlies the mined seam at distances ranging from 0.9 to 5.5 m (3 to

18 ft). The thickness of the interburden is important to the roof

control problems that develop at the mine. The mine is also

subjected to high horizontal stresses with a ratio of maximum

horizontal to vertical stresses of 1.7.

Test sites were established in tailgates of two adjacent

panels. Figure 21 shows the geologic column and an entry

cross section with the cable support design. The cables were

4.9 m (16 ft) long with four cables per row with a row spacing

of 1.5 m (5 ft). Anchorage length was 1.5 m (5 ft), leaving a

free cable length of 3.4 m (11 ft). In addition, high-capacity

dome-bearing plates as well as monster mats were installed for

surface control. At the initial test site, the interburden was 1.8

m (6 ft). Since this was the first use of cables at this operation,

a double row of cribs was installed as additional support. Even

with the crib support, cable loads averaged 98 kN (22,000 lbf),

while a total of nine cables failed in the 122-m (400-ft) test

zone. The failure was due to the large lateral movements that

occurred in the interburden. Roof separation ranged from 2.0

to 5.3 cm (0.8 to 2.1 in) and occurred between 1.2 to 1.8 m (4

to 6 ft) into the roof. Cribs in the test site were highly deformed

by the lateral roof movement. 

At the second site in the adjacent panel, interburden

thickness was 5.5 m (18 ft). The maximum increase in cable

loads was 56.9 kN (12,800 lbf) with an average increase of only

2.2 kN (500 lbf). The different sag stations showed less than

1.8 cm (0.7 in) of movement. Roof conditions remained excellent, and no roof control problems were encountered in the

entire cable section. Often the roof would remain standing one

or more crosscuts behind the face, a distance of about 45 to 90

m (150 to 300 ft).

The difference between the two sites was the interburden.

The thinner interburden consisted of weaker layers rock (stack

rock) subject to horizontal stress damage and lateral movement.

With the extensive lateral movement at the first site, a

combination of cribs and cables did maintain the tailgate.

However, the cribs did little to stop the lateral movement, and

the cables may have been able to maintain the gate road without

the cribs. With less lateral movement, the gate road probably

could have been easily maintained with the cable support. In a

third tailgate with a thin interburden and supported only by
182

Figure 20.–Two entry yield pillar tailgate showing the abutment zone outby the face, supported with cable bolts

cables and rigid trusses, large lateral movements were also

encountered. Several cable bolts did fail along with all the rigid

trusses. This occurred as the adjacent panel was mined. In this

tailgate, lateral movement of between 0.3 to 0.46 m (1 to 1.5 ft)

occurred and when the panel was mined, a roof fall did occur in

the tailgate that resulted in some delays of the longwall. Under

these very severe ground control conditions, additional support

may be required although these are tough conditions for most

support systems to control. 

CASE HISTORY 4

Case 4 is a mine located in Utah with a yield-abutment

pillar configuration [Koehler et al. 1996]. The geology of the

immediate roof consists of 0.3 to 0.6 m (1 to 2 ft) of coal and

0.3 to 0.6 m (1 to 2 ft) of mudstone overlain by a 0.3 to 0.6 m

(1 to 2 ft) layer of gray sandstone. Above this was a white

sandstone with occasional shale bands to a depth of at least 6.1

m (20 ft). The cable support design consisted of 4.3-m-(14- ft-)

long cables with four cables per row on 1.5-m (5-ft) row

spacings. The resin anchor was 1.2 m (4 ft) long, leaving a 3 m

(10 ft) length of free cable. T5 channel was used for surface

control.

The installation loads on the cables averaged 12.9 kN

(2,900 lbf). During mining of the panel, load increases on the

cables ranged from 0 to 118 kN (0 to 26,500 lbf). In the

intersections, the cable loads increased an average of 7.1 kN

(1,600 lbf). However, the highest cable loads were associated

with a near-vertical joint located near a mid-pillar instrument

site. Maximum cable load increase was 118 kN (26,500 lbf)

while the average increase was 66.7 kN (15,000 lbf). Higher

cable loads were measured along the pillar side that may be

attributable to the yield pillar and the roof breaking adjacent to

the pillar in reaction to pillar yielding. This may also explain

why the largest loads were seen at the midpillar locations. The

maximum roof movement measured in the intersection was

only 1.0 cm (0.4 in). Through the test section, there was 10 to

15 cm (4 to 6 in) of roof-to-floor convergence because of the

yield pillar. Behind the shields, the tailgate would remain open

for 15 to 41 m (50 to 135 ft). Then the entry would cave to just

behind the shields. This distance was controlled by a nearvertical joint set subparallel to the face. This was the same joint

set that resulted in the highest cable loads at the test site. Outby

the cave, the tailgate remained open with no ground control or

roof problems.

CASE HISTORY 5 

Case 5 is a mine in southern West Virginia with a threeentry abutment pillar configuration. The immediate roof at the

mine makes a transition from sandstone to shale [Mucho et al.

1996]. In some areas, the sandstone appears to be massive,

while in others it appears to be highly laminated, fossilized, and

interspersed with coal streaks. The horizontal stress is high

enough to cause damage at some locations, especially with a

thinly laminated roof. The cable system design consisted of

3.7-m (12- ft-) long cables with four cables per row on a 1.8-m

(6-ft) row spacing. These bolts were tensioned by the use of a
183

Figure 21.–Cross section of tailgate entry showing cable bolt design for case history 3.

threaded rebar head at the bottom end of the cable. The resin

anchor was 1.5 m (5 ft) long. 

When the panel was mined, less than 0.25 cm (0.1 in) of

roof separation was recorded and cable loads increased on

average only 8.9 kN (2,000 lbf). Maximum loads and roof

separations occurred as the instruments went behind the shields.

The tailgate roof area was extremely stable outby the

cave. Behind the face, the roof stayed up for a distance of 23 m

(75 ft) before the roof caved to just behind the shields. This

cyclical caving of the tailgate roof occurred throughout the test

area. In addition, when the adjacent panel was mined, there

was almost no floor heave in the cable section (tenths of inches)

as compared to the crib section where several inches occurred.
184

CABLE TRUSSES 

Cables trusses are anchored over the pillars and panels

outside the potential failure envelope and provide resistance to

roof movement along the roof line. Because cable trusses have

a high strength and flexibility and a low stiffness, they can

survive in a high-deformation and stress environment where

other supports would fail. Essentially, cable trusses move and

deform with the rock with the truss providing only limited

resistance to vertical movement [Scott 1994]. Some systems

can be pretensioned, but tensioning is probably not significant

to improved ground control, but that assures the truss is tight

when installed and therefore can respond immediately to roof

movement. Cable trusses have been used in mines since at least

the 1970s, but on a limited basis [Scott 1989; Mangelsdorf

1982]. However, in the 1990's with the advent of cable bolting

in U.S. coal mines, newly designed cable truss systems that can

be installed with roof bolters and anchored with resin grout

cartridges are now being used much more extensively as

supplemental support, especially in headgate entries. 

DESCRIPTION 

Cable trusses are constructed from a seven strand cable

usually having a diameter of 1.52 cm (0.6 in) and an ultimate

strength of 260 kN (58,600 lbf). However, cables with a

diameter of 1.27 cm (0.5 in) are also used. Cable trusses are

normally installed in a hole 3.5 cm (1-3/8 in) in diameter,

although the system can be installed in a 2.5-cm (1-in) hole.

The drill holes are typically up 2.4 m (8 ft) deep and drilled 0.6

m (2 ft) from the rib at an angle of about 45E over the coal rib.

Domed and grooved bearing plates usually 15 by 40 cm (6 by

16 in) in size are used as bearing surfaces for the rock and

cable. This allows for a two-point contact along the roof at

installation. At the drill hole-roof interface, the cable will also

be in contact with the rock, and a crushed zone may develop as

the cable loads. Cable trusses may be composed of either single

or multiple pieces, which affect how the systems are installed

but not their function.

A one-piece truss consists of a single, continuous cable

with anchorage buttons and a resin mixer on each end of the

cable in the anchor zone (figure 22) [Dolinar et al. 1996]. The

truss uses a no-spin system to mix the resin while a push button

on the cable and a special bolter wrench allow for the insertion

of the cable into the hole and through the resin with the roof

bolter. The procedure is that one end is installed, and the resin

is allowed to cure. Then the other end of the cable is placed

into the hole on the opposite side of the entry and thrust through

the resin with the roof bolter. 

With this system, installed cable truss loads ranging from

15.1 to 51.6 kN (3,400 to 8,200 lbf) have been measured. The

goal is not to develop large loads in the roof but to simply

tighten the truss so that it will provide some immediate

resistance to rock movement. 

The three-piece cable truss consists of two angle cable

bolts and a horizontal cable member (figure 23) [Oldsen et al.

1995]. The angle bolts can be constructed with nuts or

birdcages for anchorage as well as resin keepers. The cable

bolts are pushed and rotated into the hole and through the resin

using a special wrench and a roof bolter. A splice tube

assembly is attached to the angle and the horizontal cables,

which allows the pieces to be connected and the system to be

tensioned. The housing and wedge assembly that form the

cable heads are installed in the field and allow the cables to be

tensioned against the splice tube. A tensioner powered by the

hydraulics of the bolter is used to tension the system and at up

to 71.2 kN (16,000 lbf) of preload. 

ANALYSIS OF CABLE TRUSS LOADING

The loads developed in a truss can be evaluated by simple

statics. Figure 24 shows a simple free-body diagram of the

loads for a half of a truss. The following equations can be used

to describe the relationship between the reaction force R broken

into horizontal and vertical components and the cord tensions.

Yr

'T sin % (8)

Xr

' H-T cos % (9)

where T ' tension in the diagonal member,

H ' tension in the horizontal member,

Yr ' vertical reaction force, 

Xr ' horizontal reaction force,

and % ' angle of inclination of the cable.

The reaction force R may be a compilation of several

forces, especially in the case where the inclined cable bears

against the roof at the drill hole. However, these equations are

still valid for describing the vertical force Yr

 applied to the rock

by the truss or to the truss by the rock [Mangelsdorf 1979].

With a cable truss, the tension transfer between the horizontal

cord and the diagonal cords becomes more complex. Figure 25

shows a free-body diagram for the more complex loading

conditions for a cable truss. Essentially, tension load transfer

will take place by slippage of the cable over the bearing block

or plate, the bearing block or plate over the rock, or the cable
185

Figure 22.– Single piece cable truss.

Figure 23.–Three piece cable truss.

over the rock at the edge of the borehole. These load transfers

are dependent on overcoming these frictional forces. Because

of these complex loading conditions, the tension in all three legs

of the truss must be measured along with roof sag to evaluate

field performance. 
186

Figure 24.–Free body diagram of half truss showing the truss loads.

Figure 25.–Free body diagram of half cable truss, showing static loads that develop.

LABORATORY TESTS TO EVALUATE CABLE

TRUSS PERFORMANCE 

Laboratory investigations have been conducted to evaluate

the loading characteristics of cable trusses where special load

frames have been constructed to approximate the field conditions. The results from a series of tests conducted at the University of Pittsburgh indicate that only about 80% of the load is

transferred from the angle member to the horizontal member as

a result of friction across the contact blocks or bearing plates

[Mangelsdorf 1979]. In these tests, the angle member was at a

45E angle to the horizontal. When tested to failure, the load in

the diagonal cord member was 88.9 kN (20,000 lbf) where the

ultimate strength of the 1.27-cm (0.5-in) in diameter cable was

102.3 kN (23,000 lbf). The angle member achieved only 87%

of the ultimate load of the cable, with failure resulting from

bending of the cable over the contact block, which caused a

point of stress concentration and reduced the range of inelastic

deformation of the truss. The truss had reached the yield point

at the same approximate level as cable yield. At failure, the

vertical load calculated from the load measured in the cable

diagonal was 62.3 kN (14,000 lbf). The measured vertical

stiffness for a half truss was 8.8 kN/cm (2.5 tons/in ) and represents a stiffness of 17.5 kN/cm (5 tons/in) for the full truss. 

Investigators at the University of West Virginia have also

conducted laboratory tests with cable trusses in a specially

designed truss frame [Oldsen et al. 1995]. The results of these

tests were similar to those in the University of Pittsburgh study.

This study did however, provide some further insight into frictional losses and load transfer between sections of the cable and

applied loads. In these tests, a 1.52-cm (0.6-in) in diameter

cable truss with an ultimate strength of 260 kN (58,600 lbf) was
187

used, although the cables were not taken to failure. Again, the

angle between the angle and horizontal members was 45E.

However, there are differences of interpretation of the data

regarding the vertical load capabilities of the truss. In the University of West Virginia report, it is stated that the load on the

diagonal is 222 kN (50,000 lbf) when the total applied vertical

load or plate loads is 400 kN (90,000 lbf). There are two

problems with this interpretation. First, the data are being extrapolated beyond the actual test data. Second, the plate loads

are assumed to be the vertical loads. Extrapolation beyond the

test data can at times be questionable, while using loads applied

at the plates as vertical stress involves uncertainties about the

frictional conditions within both the jacks used to load the

trusses and the test frame, as well as the angle of the applied

load between the test frame and the cable. Essentially, the only

reliable measurement of vertical load should be that calculated

from the diagonal member. By using the diagonal load, the

result is that the vertical truss load is only 311 kN (70,000 lbf).

Assuming this is near cable failure, the ratio of a vertical load

of 311 kN (70,000 lbf) to an ultimate cable load of 260 kN

(58,600 lbf) is 120%. The ratio of the vertical load to the ultimate cable strength from the University of Pittsburgh tests

was 124.6:102.3 kN (28,000:23,000 lb) or 122%. These

calculations were based on symmetrical loading at the plates. 

FIELD EVALUATION OF CABLE TRUSS

PERFORMANCE

Headgate

Trusses are now being used extensively to provide supplemental support to the headgate entry where the damage to the

headgate entry is often the result of high horizontal stresses

[Mark et al. 1998; Oldsen et al. 1995]. The ability of cable

trusses to handle headgate conditions is illustrated by the

following case. 

A mine located in western Colorado had roof damage in the

headgate ahead of the face as the panel was mined [Dolinar et

al. 1996]. This was the result of horizontal stress concentration

ahead of the face and geologic features susceptible to stress

damage. (See case 1 for a more detailed description of the

geology.) A single-piece cable truss with a diameter of 1.52 cm

(0.6 in) was installed on 1.2-m (4-ft) centers in the headgate

entry (figure 26). To evaluate loading during installation and

as the panel was mined, special cable strain gages were installed

on the horizontal section of some of the trusses. The installed

load on the trusses ranged from 15.1 to 36.9 kN (3,400 to 8,200

lbf). From mining, the maximum load was 74.7 kN (16,800

lbf) for a truss just inby the face, an increase of 55.2 kN (12,400

lbf). This shows that the cable trusses were loading and resisting the roof movement. The cable trusses were able to control

the roof conditions that developed in the headgate successfully

despite the lateral movement and roof damage (figure 27). 

Other investigators have measured 17.8 to 26.7 kN (4,000 to

6,000 lbf) of increase resulting from horizontal stress damage

and cutters in headgate situations [Oldsen et al. 1997]. In these

cases, the cable trusses also successfully controlled the roof.

No cables trusses failed while the ridged trusses had.

However, if failure progresses a sufficient depth into the roof,

the dead weight load of the rock could exceed truss capacity.

This occurred at a mine in western Kentucky where cable

trusses were installed on 1.2-m (4-ft) centers [Miller 1996].

The roof failed to a rider seam when the distance to the rider

was under 3 m (10 ft) and resulted in truss failure when the

weight of the rock exceeded truss capacity.

Tailgate Support

Rigid trusses have been successfully tested as the only

secondary support in a tailgate at a test area established in a

mine in southwest Pennsylvania [Stankus et al. 1994]. In the

test, a section of tailgate entry 112.7 kN (370 ft) long was supported by trusses on 1.2-m (4-ft) centers. Loads on the

horizontal members increased by 44.5 kN (10,000 lbf) in the

abutment zone. Behind the shields, the roof did stay up for a

distance of 7.3 to 9.1 m (24 to 30 ft). 

Cable trusses in combination with cable bolts have also

been tested as the main secondary support in a section of

tailgate at another mine in southwestern Pennsylvania [Molinda

et al. 1997]. In this case, the cable trusses were installed on

2.4-m (8-ft) centers and supplemented with one row of 3.7-m-

(12- ft) long cable bolts placed along the pillar side of the entry

on 1.8-m (6-ft) centers. The tailgate outby the face and along

the shields stayed open with only minor damage to the roof

being noted. The maximum roof separation measured was just

under 2.5 cm (1 in). Behind the shields, the trusses failed

almost immediately because of the cave and only about 25% of

the entry remained open alongside the pillar for ventilation.

This small section was kept open by the cable bolts. Even this

small airway appears to have been closed off about threefourths of the way to the crosscut behind the face.

Design of Cable Truss Systems

Because cable trusses have a low vertical stiffness and are

very flexible, they can deform to the shape of the roof. This, in

combination with the high strength of the cable, makes the truss

an excellent support where especially large lateral deformation

occurs. From the laboratory tests, measured vertical truss stiffness of 17.5 kN/cm (5 tons/in) is significantly lower than the

stiffness of a cable bolt or a cable bolt system, where up to four

cable bolts would be used in place of the cable truss. 

The loading of a truss is complex; however, based on

laboratory work, the total amount of vertical load or dead

weight the cable truss can sustain appears to be about 120% of

the ultimate strength of the cable or about 311 kN (70,000 lbf)
188

Figure 26.–Cable trusses installed in the headgate entry. 

for a 1.52-cm (0.6-in) in diameter cable with symmetrical

loading at the bearing plates. Failure of the cable truss will

usually occur between the anchor hole and the bearing plate

when cable tension load in the angle section is around 87% of

the ultimate strength of the cable (figure 28 ) [Tadolini et al.

1998]. Furthermore, a cable truss in situ can be subject to

asymmetrical loading, resulting in an even lower load capacity.

Therefore, to determine the performance of a cable truss in the

field, it would be necessary to measure the strains or loads on

all three sections of the cable as well as measuring roof sag

[Mangelsdorf 1979]. For cable trusses in general, strain measurements are usually determined only on the horizontal

member. Thus, a complete picture on the performance of cable

trusses in situ has not been obtained. 

Generally, trusses are installed between the existing rows

of primary roof support, so spacings will be on 1.2-or 1.5-m (4-

or 5-ft) centers. However, if the roof failure is deep enough, the

dead weight load of material can exceed truss capacity. Based

on laboratory tests, for a 1.52- cm-(0.6-in) long cable with an

ultimate strength of 260 kN (58,600 lbf) and the truss carrying

a load 120% of cable strength, the dead weight load capacity is

313 kN (70,500 lbf). With 5.5-m (18-ft) wide opening, a 1.2-m

(4-ft) truss spacing, and a rock density of 2,307 kg/m3

(144 lbf/ft3), a failure depth of about 2.1 m (7 ft) would exceed

this capacity. In such cases, either tighter truss spacing, higher

capacity trusses, or additional supplemental support must be

used in conjunction with the truss system.

For the cable trusses, anchorage requirements are the same

as for a cable bolt where a minimum of 1.2 to 1.5 m (4 to 5 ft) of

anchorage length should be used. The trusses are anchored in

angle holes that are usually drilled at a 45E angle over the coal

rib. Other angles can be used, but this will affect loading and

load distribution in the truss. These angle holes allow the truss

anchorage to be outside the potential failure zone. Once the

anchorage is undercut so that when the trusses are behind the

face, this is no longer the situation and the truss fails because of

the loss of the anchor. Similar conditions may also develop in

intersections or the shield zone where the angle member is not

anchored above a coal rib, but in a potential failure zone.

Bearing plates are used and installed up to around 15 to

30 cm ( 6 to 12 in) from the anchor hole. These bearing plates

allow two points of contact on the roof, lessens the cable bend,

and allows for more efficient load transfer along the cable. 

Cable trusses have been used successfully as supplemental

support in the headgate to control damage from high horizontal

stresses that can develop near the longwall face. The strength,

low stiffness, and flexibility of the cable trusses are important

characteristics that allow the support to survive and maintain

control of a damaged and highly deformed roof. Other types of

support, especially rigid trusses, have failed under conditions

where large lateral movements occur. As the main secondary

support in the tailgate, cable trusses have been relatively

successful in a few test cases. However, the trusses do have

trouble maintaining the tailgate open behind the shields because

of loss of anchorage. Also, there are no data to indicate

whether there is any loss of anchorage and therefore support as

the cables are undermined between the face and the back of the

shield. When mining through an intersection, either side of the

truss could fail as a result of loss of anchorage because the

truss is anchored in a potential failure zone.
189

Figure 27.–Roof damage from horizontal stress around cable

truss in headgate just outby the face.

Figure 28.–Shear failure of cable truss near bearing plate.

SUMMARY AND CONCLUSIONS 

Cable technology as used in the longwall gate roads in U.S.

coal mines was developed in the 1990's. This technology

includes cable bolts and cable trusses. Cable bolts consist of a

headed cable utilizing a partial grout column anchor formed

from a resin cartridge and installed with a roof bolting machine.

When evaluating cable bolts, there are several characteristics and components of the cable system that are important

to bolt performance. This includes cable strength, elongation,

stiffness, and shear resistance, and system anchor capacity. In

general, the cable should be the weakest part of the system.

Therefore, to exceed the ultimate capacity of the cable, the

anchorage length should be a minimum of 1.2 to 1.5 m (4 to

5 ft) long. The stiffness of the cable system, the ability to resist

loading, is determined by the free length and elastic properties

of the cable. As determined from in situ pull tests, the elastic

modulus of the cable was found to be about 131.7 GPa

(19.1 million lbf/in2

). This value can be used to calculate cable

bolt stiffness. For improved long-term performance, galvanized

wire strands or epoxy-coated cable should be used to resist

cable corrosion and limit the potential for any strength

reduction of the cable. Furthermore, high- capacity bearing

plates and heavy-duty mats or channel provide added protection

with surface control and an element of structural support for the

immediate roof.

Design of cable bolt systems as secondary support in tailgate entries is based to a large extent on suspension, although

this is somewhat of an oversimplification of the conditions that

can develop. Cable lengths are determined by the depth in the

roof of a potential failure horizon over the entry plus an adequate anchorage length. The number or density of cables will

then be determined by the dead weight load of rock below that

failure horizon. Lateral roof movement may also cause significant loads to develop in the cables where the cables resist the

movement and increase the residual shear strength of the rock

However, in some cases, the cables may not be able to stop or

limit lateral movement and, as a result, can fail. It has been

estimated that the cables can handle up to 5 to 10 cm (2 to 4 in)

of lateral movement. There are measures that can be taken to

reduce the potential for cable failure, including locating the

bolts outside the highest lateral deformation zones or using

yielding bolt heads. 

For supporting tailgates, there are three zones that must be

considered when evaluating the design and performance of the

cable system. These zones include the outby abutment zone for

both vertical and horizontal stress, the shield zone, and the cave

zone. In situ tests have been used to further define and confirm

cable bolt designs and performance in each of these zones. In

these test cases, the number of cables used per row was four

with 1.2- to 1.8-m (4-to 6-ft) row spacings. The cable lengths

at the site varied from 3.7 to 4.9 m (12 to 16 ft). These cable

bolt systems were very successful in supporting longwall

tailgate entries with few resulting ground control problems.

From a ground control standpoint, cable bolts have an

advantage over standing support where they do not resist main

roof-to-floor convergence. This is especially important with a

yield pillar system because much of the capacity of the standing

support will be taken up by this convergence. Although cable

bolts have maintained the tailgate entry behind the shield for

long distances, the cave and roof geology are the main factors

that determine this distance and not the cable system design.

Therefore, cable bolts cannot guarantee that the tailgate can be

kept open to the first crosscut behind the face for ventilation.

Cable trusses were greatly improved in conjunction with

cable bolt development, and as a result, are now used more

extensively than previously, especially as supplemental support

in headgate entries. In the headgate, the cable truss has been used

to control damage caused by horizontal stress. High strength and

flexibility and low stiffness are reasons why trusses can survive

in a high-stress and high-deformation environment and still

function to maintain a highly deformed roof.
190

Furthermore, the cable truss anchorage is outside the potential

roof failure zone. Therefore, cable trusses have been successful

in providing supplemental support in critical headgate entries.

However, based on laboratory tests, the capacity of trusses

to carry dead weight loads appears to be only about 120% of the

ultimate strength of the cable. More tests, including in situ

studies, are required to determine if this capacity could be used

for design or must be modified. 

To evaluate the performance and capacity of a truss in situ

though requires monitoring loads on all three cable legs along

with roof sag. Although trusses have been instrumented, to date

this has not been done to the level required for a complete

evaluation of their performance. As secondary support in the

tailgate entry, cable trusses have been tested or used only on a

limited basis. Behind the shields, trusses can only keep the tailgate open for a very limited distance. Beyond such distances,

there are questions on how well heavily loaded trusses will

perform when the anchors are undercut by mining or in intersections where anchors are not supported by a coal rib. 

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193

MATERIAL HANDLING CONSIDERATIONS FOR SECONDARY ROOF

SUPPORT SYSTEMS

 By Thomas M. Barczak1

ABSTRACT

Secondary roof support systems play a vital role in preserving the safety of underground mine workers by

preventing the unintentional collapse of the mine roof. Hundreds of thousands of standing roof supports are

constructed each year in underground coal mines. Historically, wood and concrete cribs and timber posts have

been used for secondary roof support. These support constructions require the handling of heavy and bulky

materials, causing numerous injuries to the mine workers and resulting in more than 40,000 lost workdays in

the past 9 years. Since 1993, various alternative support technologies have been developed. These new

support technologies not only provide superior roof support, but many also have significant material handling

advantages by using smaller and lighter weight materials, fewer components, mechanically installed support

systems, and pumpable support systems. The National Institute for Occupational Safety and Health (NIOSH)

has conducted extensive material handling research and has developed recommended lifting thresholds to

reduce material handling injuries. An analysis of roof support construction reveals that conventional support

materials used in wood and concrete cribbing exceed the recommended lifting thresholds, while the engineered

support systems are closer to the recommended weight thresholds. Finally, recommendations are made relative

to proper lifting techniques and material handling practices to prevent injury to mine workers when

constructing roof supports in underground coal mines.

1

Research physicist, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh PA.
194

Figure 1.–Number of timber-handling accidents.

Figure 2.–Incident rate for timber-handling injuries.

INTRODUCTION

The primary goal of roof support is to prevent unplanned

roof falls from occurring. While falls of ground are the leading

cause of fatalities in underground coal mining, many more

injuries from the effort required to install these essential roof

support systems than from the fall of ground due to poor roof

support design. There is also continued pressure to install supports more quickly to keep pace with the escalating productivity, particularly in modern longwall mining operations. These

requirements, coupled with the fact that the workforce is aging,

strongly suggest that a higher priority should be given to the

material handling considerations and construction practices during the roof support selection process.

During the 1990s, there was an unprecedented increase in

the development of standing roof support systems to replace

conventional wood and concrete cribbing and timber posts that

have historically been used for secondary roof support

throughout the mine. These innovative roof support systems

were designed to provide superior roof control, but most also

provide material handling benefits that allow supports to be

constructed with less effort and at faster rates.

This paper highlights the material handling characteristics of

these modern roof support systems and describes their impact

on the installation of secondary roof support and safety of the

mine workers performing this function. A complete assessment

of the transportation and construction requirements for each

support is made to provide mine operators with a guide for

evaluating these alternative support technologies. Regardless

of the physical properties of the support, injuries can also be

prevented by using proper lifting techniques and avoiding excessive stress that occurs when recommended lifting thresholds

are exceeded. These issues are also addressed through some

practical examples.

INJURY INFORMATION

According to the Mine Safety and Health Administration

(MSHA) accident database, over 55% of the permanent disabling

injuries in underground coal mining during 1992-94 were due to

material handling [MSHA 1996]. An even greater number of

nondisabling injuries may be linked to material handling in

mining. It is likely that these accident trends will continue and

perhaps grow worse due to an aging workforce. It is also very

likely that younger and inexperienced miners that will replace

retiring miners will also experience a high incidence rate for these

kind of accidents until they become more skilled. 

The construction of secondary support systems has historically required repetitive lifting of large volumes of bulky

support materials. Coupled with the poor conditions, underground limited space and maneuverability, and the fact that

many of the support materials exceed 40 lb, this activity is

responsible for numerous injuries to mine workers. Such

activity has been classified by the National Occupational Health

Survey of Mining as a heavy lifting risk factor, exposing miners

performing this activity to a high risk for musculoskeletal

repetitive trauma disorders [Winn and Biersner 1996]. Thus,

material handling should also be a primary consideration in the

application of longwall tailgate support technologies.

A review of the MSHA accident database reveals that 1,483

lost-time accidents were reported from 1990 to 1998 associated

with timber handling. Further review of these accidents indicates

that 1,204 were directly related to support construction,

accounting for 40,147 lost workdays during this 9-year period.

The average lost time per incident was 33.34 days. Figure 1

shows the number of material handling accidents attributed to crib

construction from 1990 to 1998; figure 2 shows the incidence rate

for the same period. Generally, both the number of accidents as

well as the incidence rate dropped during this period, except for

a moderate increase in 1998. Without more extensive data than

are available in the current database, there is no apparent reason

for the increase in 1998. It could be that more attention has been

drawn to material handling issues recently, and minor incidents

that were previously not reported or mislabeled are now being

reported. It might also be a 1-year anomaly.
195

Figure 3.–Comparison of support technologies used in 1993

and 1996.

Figure 4.–A, Comparison of timber-handling incident rate with

the implementation of alternative engineered timber support

systems. B, Comparison of days lost due to timber handling with

the implementation of alternative engineered timber support

systems.

Figure 5.–Assessment of lost workdays

relative to body part for timber-handling injuries.

Figure 6.–Bodily assessment of lost workdays per incident for

timber handling accidents.

The decrease in injuries from roof support construction

correlates with the implementation of new support technologies

that exhibit material handling advantages over conventional wood

cribbing. These alternative support systems are addressed in

detail in the remainder of the paper. Before 1992, all longwall

tailgates used either conventional wood or concrete cribbing. As

shown in figure 3, only 39% of the longwall tailgates were

supported with conventional wood or concrete cribbing in 1996.

Figure 4A compares the incidence rate for timber material

handling injuries with the replacement of conventional cribbing

with engineered timber supports. As seen, the trend of

decreasing material accidents continues during the period when

new support technologies were introduced. Figure 4B shows

that the severity of injury decreased considerably after 1994,

correlating with the implementation of the engineered alternative support systems. These findings suggest that these lighter

weight support materials are having a positive impact on reducing material injuries associated with support construction in

underground coal mines.

Figure 5 shows a breakdown of lost workdays due to timber

handling injuries in underground coal mines. As might be expected, 71% of the lost workdays are due to back-related

injuries. Figure 6 depicts the number of lost workdays per
196

Figure 7.–Construction of four-point and nine-point

conventional wood crib supports.

incident. Here it is seen that the most severe injuries are to the

trunk of the body (80 days per incident). The next most severe

injury is to the neck (57 days per incident), followed by back

injuries at 40 days per incident. In conclusion, timber handling

activities can result in injury to several parts of the body, and

many injuries will be severe enough to cause several days to

several months of lost work time.

SUPPORT DESIGN IMPROVEMENTS THAT REDUCE MATERIAL HANDLING

REQUIREMENTS

Conventional wood and concrete cribs require piecemeal

construction of relatively heavy and bulky materials weighing

35 to 55 lb. Significant material handling advantages are realized in various alternative support technologies by the use of

lighter weight and more compact support components; other

support systems use fewer components in the support construction. Manual material handling is all but eliminated in

supports that are installed with a machine. Innovative supports

that are pumped in place in the mine greatly reduce transportation requirements and minimize manual effort in the

confined space of the underground mine where material

handling efforts are considerably more difficult than on the

surface.

SMALLER AND LIGHTER WEIGHT COMPONENTS

Conventional wood cribs are typically constructed in fourpoint or nine-point configurations, as shown in figure 7. In

these configurations, the roof load is carried only through the

area where the adjacent timbers contact one another. For

example, in a four-point wood crib, this occurs only at the four

corners of the support structure. As seen in figure 7, over 50%

of the timber does not contribute to the capacity of the support.

Modern engineered timber supports, such as the Link-N-Lock

crib and the Link-N-X crib developed by Strata Products USA

and the Tri-Log crib developed by American Commercial, Inc.,

as shown in figure 8, achieve full timber contact by notching

the timbers. By establishing full timber contact, the timber

dimensions can be reduced without sacrificing supporting

capability. Table 1 compares the timber dimensions and

weights for conventional wood cribs and selected engineered

timber supports.

As shown in table 1, the timber weight is reduced by about

a factor of about 3 for selected designs of the Link-N-Lock, TriLog, and Link-N-X cribs. In addition, these designs provide

1.6 to two times the support capacity of a conventional fourpoint oak crib. Comparing these engineered timber supports to

a four-point crib constructed from poplar timbers, which most

closely represents that of a mixed hardwood crib, it is

determined that despite half the timber weight, the support

capacity is increased by a factor of 2.4 to 3.0 for the engineered

crib support structures.

CONSISTENT MATERIALS

Studies have shown that one of the risk factors for back

injuries is unexpected or unanticipated movements [Marras et

al. 1987]. When materials are of different and unknown

weights, there is an increased chance of back injury. Thus, a

miner constructing conventional wood crib supports from

timbers consisting of mixed wood species where one piece of

wood may weigh 30 lb and the next piece may weigh 50 lb is

more prone to injury than if the timbers are all of the same

weight. Engineered timber supports such as the Link-N-Lock,

Link-N-X, Hercules, and Tri-Log cribs all use the same species

of timber for the support construction. Thus, these timbers are

much closer in weight than the mixed wood species used in

conventional wood crib supports. This reduces the injury potential that occurs when a miner is expecting a light piece of

wood, but instead gets a much heavier piece of wood during the

support construction process.

FEWER COMPONENTS

The Hercules crib is an example of a support that is designed

to provide superior support capability with fewer pieces required for the support construction. It is constructed from preformed mats that are stacked on top of one another (figure 9).

The Hercules crib can be constructed in a variety of configurations to provide a wide range of support designs and
197

Figure 8.–A, Link-N-Lock crib; B, Tri-Log crib, and C, Link-N-X crib.

Table 1.—Comparison of conventional wood cribbing to engineered crib support systems

Parameter Four-point 

oak crib

Link-N-Lock 

(24-in)

Tri-Log 

(30-in)

Link-N-X

(30-in)

Four-point

poplar crib

Timber length, in . . . . . . . . . . . . . . . . . . . . . 36 24 30 30 36

Timber width, in . . . . . . . . . . . . . . . . . . . . . . 6 3.5 4 6 6

Timber rise, in . . . . . . . . . . . . . . . . . . . . . . . 6 6 6 4 6

Timber weight, lb . . . . . . . . . . . . . . . . . . . . . 40.5 13.1 19.1 19.1 31.6

Support weight per foot of height, lb/ft . . . . . 208 102 88 90 126

Capacity at 2 in of convergence, tons . . . . . 52 94 105 85 35

performance profiles. The comparison in table 2 shows the

HM-9(308) provides 38% more capacity than a four-point oak

crib. The HM-9(308) mat weighs slightly more than a 6- by 6-

by 36-in oak timber, but since each mat provides a full layer in

the Hercules support, the material weight per foot is 18% less

than a four-point crib.

There is also a variety of prop-type supports that are

constructed as units rather than the piecemeal construction

required for crib-type supports (figure 10). The unit weights of

these supports are typically greater than the unit weights of the

piecemeal crib components, but since they are installed as unit

and generally stood in place as opposed to being lifted, the

cumulative effort to install the support is considerably less. In

addition, the construction of the prop-type supports does not

require lifting of material above the shoulder and thus reduces

the risk of injury by eliminating this awkward lifting condition.

The primary advantages of the prop supports from a material

handling perspective are twofold: (1) supply cars can transport

more support units and (2) supports can be installed in less time

than crib-type supports since there are considerably fewer

pieces to handle. Table 3 documents transport volumes and

installation rates of various prop-type supports in comparison

to conventional wood cribbing.

MECHANICALLY INSTALLED SUPPORT SYSTEMS

Other support systems are installed with machines that all

but eliminate the material handling effort by the miners. Table

4 describes relevant material handling parameters for three

mechanically installed support systems. The Burrell Can developed by Burrell Mining Products pioneered the development

of mechanically installed supports. The Burrell Can is installed

with a hydraulically powered gripper claw that can be attached

to a loading machine or a scoop (figure 11A). Strata Products

has recently developed the Star Prop that can be installed with

the aid of the Prop Handler or the Microtrax. Figure 11B shows

the installation of Propsetter supports using these mechanical

aids.

Table 2.—Comparison of a Hercules crib to conventional

four-point oak crib

Parameter

Nine-point 

oak crib

(6x6x36 in

timbers)

Hercules crib

HM-9(308)

Timber weight, lb . . . . . . . . . 40 44

Support weight per foot 

 of height, lb/ft . . . . . . . . . . 162 132

Capacity at 2 in of 

 convergence, tons . . . . . . 52 72

Table 3.—Comparison of prop-type supports to conventional

four-point wood crib

 Parameter

Four-point

 crib Propsetter ACS

Yippi

prop

Stretch

prop

Transport

 volume, ft3 . . . . 24 4.6 4.3 2.8 3.7

Supports per

 supply car . . . . 16 84 91 140 107

Installation rate,

 supports per

 shift . . . . . . . . . 15 48 53 80 60
198

Figure 9.–Hercules crib.

Figure 10.–A, Propsetter; B, ACS; C, Yippi support; D, Stretch prop.

Figure 11.–A, Installation of the Burrell Can; B, installation of the Propsetter support.

Table 4.—Comparison of some mechanically installed support

systems

Parameter Burrell Can

(24-in diam)

Propsetter

(10-in

diam)

Star Prop

(standard

design)

Installation rate, supports

 per shift . . . . . . . . . . . . 1

40 50 50

Transport numbers, sup-

 ports per car . . . . . . . . . 12 65 43

Capacity at 2 in of con-

 vergence, tons . . . . . . . 85 68 85

 1

The installation rate of the Burrell Can support can vary considerably,

depending on the availability of equipment for delivery of the supports

and timbers for topping off the supports. The 40 supports per shift is a

representative installation rate for Burrell Can supports that includes

delivery time as part of the installation process in a well-planned

operation; however, this rate can range from 30 to 50 supports per shift.
199

Figure 12.–A, Surface pumping station for Pumpable Crib; B, Pumpable Crib; and C, underground installation of

Pumpable Crib in longwall tailgate.

PUMPABLE ROOF SUPPORT SYSTEMS

Another technology that offers material handling advantages

is the use of pumpable support systems where a cementitious

grout is pumped in place into some form in the mine entry. This

process greatly reduces the transportation effort and can reduce

the material handling effort with support installation. Both

Heintzmann Corp. (figure 12) and Fosroc Corp. offer pumpable

support systems. The Stretch Prop (figure 10D) developed by

Ferrocraft, Inc., is a prop-type support that uses a pumpable

cementitious grout to extend and fill the support during the

installation process. Table 5 compares the Heintzmann Corp.'s

pumpable crib to a conventional four-point oak crib.

Table 5.—Comparison of pumpable crib and conventional

Four-point oak crib

Parameter four-point

oak crib

Pumpable

crib

Supports per supply car . . . . . . . . . . . 16 100

Construction work, ft-lb . . . . . . . . . . . . 5,838 900

Installation rate, supports per shift . . . 15 1

50

Capacity at 2 in of convergence, tons 52 240

 1

The installation of the pumpable crib in this example is based on

a surface pumping operation and a underground crew using a total of

7 people. The installation rates may vary depending on the crew size

and the pumping requirements, but the 50 supports per shift is

considered to be representative of a typical operation.

RECOMMENDED LIFTING THRESHOLDS TO REDUCE MATERIAL HANDLING INJURIES

Low-back pain and associated injuries from lifting are the

most costly occupational health problems facing our Nation

[NIOSH 1997]. As a result, the National Institute for

Occupational Safety and Health (NIOSH) has conducted

extensive ergonomic research to evaluate lifting mechanics and

their impact on the human body. Through this research, a lifting equation has been developed that determines a recommended weight limit (RWL) for lifting in various postures

[Waters et al. 1993]. The RWL is defined for a specific set of

task conditions as the weight of the load that nearly all healthy

workers could perform over a substantial period of time (e.g.,

up to 8 hr) without an increased risk of developing liftingrelated lower back pain.

RWL ' LC × HM × VM × DM × AM × FM × CM,

where RWL ' recommended weight limit;

LC ' load constant ' 51 lb; and HM, VM, DM, AM,

 FM, and CM are various multipliers.

The lifting equation begins with a load constant (LC) of 51

lb, which is the maximum recommended lifting weight under

ideal conditions. Interpreted as the most optimal conditions, no

more than 51 lb should be lifted regardless of the task

characteristics. Putting this into perspective relative to support

construction, a 6- by 6- by 45-in oak timber that has been dried

for about 30 days weighs 52 lb, whereas a green 6- by 6- by 36-

in oak timber weighs as much as 52 lb. The 51-lb upper limit

is then decreased through six multiplicative coefficients that

further define the lifting task. The variables that impact the

recommended weight limit are defined as follows:

1. Horizontal Position (HM): The horizontal position refers

to the horizontal distance of the lifted object (where the person

holds the object) from the centerline of the person's feet or,

more specifically, the ankles (figures 13-15). The recommended weight limit is reduced as the object's distance from the

body increases.

2. Vertical Position (VM): The vertical position refers to the

distance the object is from the floor level before lifting
200

Figure 13.–Biomechanics of lifting showing effect of horizontal distance on muscle force.

Figure 14.–Material should be kept close to the

body (A) as the material is held away from the body (B)

the risk of injury increases.

Figure 15.–Picking up a 40-lb Hercules mat. Holding the

mat so that it is positioned close to the body (A) reduces the

lifting effort compared to lifting the mat in a horizontal position

(B) where the center of gravity is farther from the body.

Figure 16.–Avoid twisting.

is executed. The optimum distance is 30 in for an average

worker who is 5 ft 6 in tall [Waters et al. 1994]. The vertical

position multiplier decreases linearly with an increase or

decrease in height from the optimal 30-in position.

3. Lifting Distance (DM): The vertical lifting distance is

how high the object is lifted. The recommended weight limit is

reduced as the lifting height increases. The multiplier ranges

from 1.0 for a height of 10 in to 0.85 for a maximum height of

70 in.

4. Twisting Factor (AM): The twisting factor is referred to

as the asymmetric component, which is defined as the rotation

(twisting) of the body required to place the object at a destination (figure 16). It is expressed in angular degrees relative

to the feet position at the origination of the lift and the twisting

of the body at the lift destination. The recommended weight

limit can be decreased significantly for asymmetric lifting

compared to symmetric lifting.

5. Frequency Factor (FM): The lifting frequency refers to

the average number of lifts in typical 15-min period. In addition

to the number of lifts, the lifting frequency multiplier is influenced by the duration of the lift and the vertical position of the

hands at the origin of the lift. The duration of the lift is classified into three categories: (1) short, (2) moderate, and (3) long.
201

These categories are based on the work/rest pattern of the job.

Short duration defines lifting tasks that have a duration of 1 hr

or less followed by a recovery (nonlifting) time equal to 1.2

times the work time. Moderate duration includes lifting periods

of 1 to 2 hr followed by a recovery time of 0.3 times the work

time. Long duration is defined as lifting periods between 2 and

8 hr, with standard industrial rests allowances (morning, lunch,

and afternoon breaks). As one would expect, the lifting limits

are reduced as the frequency of lifting and/or the lifting period

increase.

6. Handling Factor (CM): The handling factor technically

referred to as the "coupling component" attempts to evaluate the

worker's gripping method used on the object. It also considers

the vertical location of the hands throughout the entire range of

the lift. The coupling component is also divided into three categories: (1) poor, (2) fair, and (3) good. The coupling multiplier

ranges from 0.9 to 1.0.

Some representative support construction examples are shown

in table 6. Analyzing these data, it can be shown that the lifting

threshold is most sensitive to how close the support material is

held to the body during the lifting process (horizontal factor).

The recommended lifting weight is reduced from 42 to 14 lb

when the horizontal distance increases from 0 to 12 in with an

initial horizontal origin 6 in from the body (feet) centerline. From

a biomechanics analysis as shown in figure 13, it is apparent that

muscle force greatly increases as the object is moved farther from

the spine. The lifting threshold is also reduced when the body

must turn to position the support material during the lifting

process. For example, with zero horizontal lifting the recommended lifting weight decreases from 42 to 36 lb when the

angular rotation changes from 0E to 45E. Twisting causes increased stresses in the intervertebral disks in the spinal column

[Gallagher et al. 1990]. Excessive twisting may cause some of the

disk fibers to break, which severely weakens the disk. It is the

opinion of some researchers that back injuries caused by twisting

are much less likely to heal than injuries due to simple bending

[Gracovetsky and Farfan 1986].

The reduced material handling requirements for the engineered timber supports are clearly seen when these factors are

considered. Reviewing a previous example, a typical wood crib

is constructed from oak timbers weighing 40 lb while the engineered Link-N-Lock (24 in) and Tri-Log cribs (30 in) are

constructed from timbers that weigh 13.1 and 19.1 lb, respectively. Therefore, while the standard crib exceeds the

recommended lifting thresholds in all but ideal conditions, the

Link-N-Lock and Tri-Lock crib meet the typical and extreme

conditions. Thus, significantly less musculoskeletal injuries

would be expected from the construction of these engineered

timber supports compared to conventional wood cribbing.

Table 6.—Recommended lifting weights for selected task parameters relevant to roof support construction

Lifting

height, in

Horizontal hand position from

centerline of feet, in Lifting

frequency,

lifts/min

Asymmetric

angle, E

Recommended

liftingweight, lb Origin Destination

60 . . . . . . . . . . . . . . . . 6 6 1 0 42

60 . . . . . . . . . . . . . . . . 6 12 1 0 21

60 . . . . . . . . . . . . . . . . 6 18 1 0 14

60 . . . . . . . . . . . . . . . . 12 12 1 0 21

60 . . . . . . . . . . . . . . . . 12 18 1 0 14

60 . . . . . . . . . . . . . . . . 18 18 1 0 14

60 . . . . . . . . . . . . . . . . 6 6 1 30 38

60 . . . . . . . . . . . . . . . . 6 6 1 45 36

60 . . . . . . . . . . . . . . . . 12 12 1 30 19

60 . . . . . . . . . . . . . . . . 12 12 1 45 18

60 . . . . . . . . . . . . . . . . 18 18 1 30 13

60 . . . . . . . . . . . . . . . . 18 18 1 45 12

INJURY PREVENTION THROUGH PRACTICAL LIFTING TECHNIQUES

AND MATERIAL HANDLING PRACTICES

Miners constructing roof supports in an underground mine

environment are at risk of musculoskeletal injuries. As previously described, these injuries can be minor or severe enough

to cause permanent disability. Although NIOSH is conducting

research to promote the development and application of

lightweight materials and mechanical aids to reduce the effort

required for support construction, injuries can also be prevented

by following proper lifting techniques and support construction

practices. Following are several practical recommendations to

this effect.

1. Hold the support material as close to the body as

possible.

As discussed previously, the recommended lifting weights

decrease significantly when the lifted object is moved away

from the body. Thus, material handling practices that facilitate

keeping the material close to the body, as shown in figure 14A,

should be seriously pursued. Handling material in the manner

shown in figure 14B should be avoided, particularly when the

material weight exceeds 20 lb. Another example is shown in
202

figure 15 with the Hercules crib mat. Ideally, two people

should lift the mat. If this is not possible, then lifting the mat in

a vertical orientation is better than in the horizontal position.

Another good practice is to position oneself as close as possible

to the destination of the support component during the lifting

process. For example, when constructing a crib, it is better to

walk to the side of the crib where the timber is being installed

than to reach across the support structure to lift the crib block

in place. Although this may take more time, it significantly

reduces the stress on the back and, overall, will reduce fatigue

so that longer construction times can be realized.

2. Avoid lifting above the shoulders.

When possible use a ladder or some other device to increase

the standing height as opposed to lifting above the shoulder.

Lifting effort increases significantly once the object is lifted

above the waist, because the arms are now required to

accomplish the lifting without any further benefit from the leg

muscles. Lifting above the shoulder requires more energy and

can create an additional risk factor due to awkward lifting

where the arms might need to be extended or the body twisted.

Lifting materials above the shoulders may also jeopardize the

miner's balance, creating an unstable posture that increases the

risk of falling and causing further injury.

3. Avoid lifting from the floor.

As much as possible, have support material stacked and

delivered to the work site on pallets to minimize material lifting

from the floor level. Often in underground mines, the packaging is unnecessarily destroyed during the transportation process. The lifting equation research suggests that a 30-in starting

height is the most efficient for an average height person. As a

rule, any material handling task should avoid lifts and placements below the knuckle (measured from a relaxed standing

posture) and above the shoulder.

4. Lift in one smooth operation.

A large number of back injuries are attributed to sudden or

unexpected movements. The back stress incurred by the worker

in this situation is often two to three times as great as when the

load is expected [Marras et al. 1987].

5. Avoid excessive twisting during the lifting process.

When stacking support materials such as in building a crib,

there is a natural tendency to keep the feet still and twist the

body to minimize the support construction time (figure 16).

However, this practice causes excessive strain on back

muscles and vertebrae and should be avoided, particularly

when lifting materials in excess of 35-40 lb. Repositioning

the body to remain directly in front of the lifting destination

will significantly reduce the likelihood of severe back injury

that can result from twisting motions.

6. Use mechanical assists whenever possible.

Several examples of support systems that are designed for

installation with mechanical assists were previously described.

The use of these systems and the development of others should

be encouraged. Figure 17 depicts one apparatus used for setting

timber posts.

7. Rest when needed.

The basic premise of this recommendation is that muscle

fatigue can lead to or increase the likelihood of musculoskeletal

injury. As the metabolic demands associated with support construction increase, more frequent rest periods are required. The

Work Practices Guide for Manual Lifting [NIOSH 1981]

recommends that for occasional lifting, the metabolic energy

expenditure rate should not exceed 9 kcal/min for physically fit

males and 6.5 kcal/min for physically fit females. Energy

expenditure in crib building has been measured in one study at

8.48 kcal/min [Gallagher 1987]. Thus, crib building approaches

the recommended energy threshold for a physically fit male.

The American Industrial Hygiene Association has developed

algorithms to determine the recommended rest interval for various work-related tasks. The rest break will allow the heart rate

and breathing rate to return to normal, as well as allowing the

metabolic end products of the muscle exertion to dissipate and

reoxygenate the muscle. The work rest cycles are based on the

energy expenditure required to perform the task. Using the 8.48

kcal/min for conventional wood crib, the recommended rest

interval equates to 75% of the work time [Gallagher 1987].

This means that if a crib crew spent 30 min constructing a crib,

the recommended rest time is 22 min before building the next

crib to prevent excessive muscle fatigue.

8. In low-seam heights, lift from a stooped position versus

a kneeling position.

Studies have shown that miners have a significantly lower

lifting capacity, 10 lb over on average, in the kneeling posture

than in the stooped posture [Gallagher et al. 1990]. In terms of

biomechanics, the large muscles of the lower back contract

much more vigorously in the kneeling posture than in the

stooped posture. This implies a greater compressive loading of

the spine when kneeling, thus a greater chance of back injury.

However, this recommendation needs to be qualified in the

sense that if the load can be placed between the legs when

squatting, then the load is closer to the spine, causing less stress

on the back.
203

Figure 17.–Apparatus used for setting timber posts. (Photo

courtesy of Strata Products USA).

9. Get help when the required lifting load exceeds the

recommended weight limit.

It is important to remember that the recommended weight

limit is significantly less than a person's maximum lifting

capacity. The intent of the recommended weight limit is to define thresholds that will prevent injury. By asking for help, the

probability of back injury can be significantly reduced when the

load is shared among two or more people.

10. Promote your own physical fitness.

Support constructing can be a very demanding job. Studies

have shown that many back injuries may be prevented by

strengthening the lower back and abdominal muscles. Stretching before lifting can also be very important in preventing back

strains.

SUPPORT MATERIAL HANDLING PRODUCT GUIDE

Material handling requirements can be divided into two

functions: (1) delivery of material into the mine and (2) construction of the roof supports. A summary assessment of various support technologies for these material handling functions

is provided in tables 7 and 8.

Support materials are typically loaded onto either supply cars

or shield carriers and transported into the mine by a rail haulage

system. A typical supply car will be 7 ft wide and 20-24 ft long.

The car volume then depends on the bed height above the track

rail, which is determined by the seam height. The shield cars

are smaller (14-16 ft long), but are 8-12 in lower, which allows

more clearance to the mine roof. Thus, a higher stack of material can be transported on them. At the mine entry, the supports are typically unloaded by a diesel- or battery-powered

forklift, bucket scoop, or shield hauler. How the supports are

packaged is critical to the overall material handling requirements. Generally, the goal is to bundle and unload the materials

in full support increments. This will minimize the construction

effort by placing the correct amount of material in the vicinity

where the support is to be constructed and avoid the extra and

generally manual effort in carrying surplus materials to a new

location. Table 8 documents the relevant parameters associated

with transportation of support material into the mine.

Table 7 shows that prop-type supports (Propsetter, Star

Props, Lock-N-Load Props, Alternative Crib Support (ACS),

and Yippi Prop) and the pumpable supports in collapsible

containers (Pumpable Crib and Tekcrib) provide significant

transportation advantages in terms of reduced volumes compared to crib-type supports that are constructed in piecemeal

fashion or large-diameter, precast, concrete support structures.

The next issue pertains to the construction of the supports.

The primary factor to consider is the labor involved in the

support construction and the rate of installation of the support

structures. From an injury perspective, the size, weight,

handling method, and the number of pieces handled per unit

time are relevant factors are previously addressed. The energy

required to construct the support would be a useful method to

define the risk assessment for musculoskeletal injuries. However, an energy analysis that considers the physiological demand on the human body is quite complex and is beyond the

scope of this paper. Instead of this complex analysis, a computation of work (weight of piece times the lifting height) can be

done. The efficiency of the support construction can be judged

in terms of the installation rate and the construction effort

(work). Table 8 compares the relevant parameters for construction of the various support technologies.

Table 8 shows that the construction effect increases dramatically for supports that require piecemeal construction using

large numbers of heavy pieces such as conventional concrete

cribbing. The work required for constructing a steel-fiberreinforced donut concrete crib is 40% more than the work

required for a conventional four-point wood crib. As a result,

the installation rate for these supports are among the lowest of

all support technologies. The benefits of the engineered crib

supports (Link-N-Lock, Link-N-X, and Tri-Log cribs) are

clearly seen in comparing the construction work to that of

conventional wood cribbing. The least effort is required for the

construction of the Propsetter, ACS, and the Burrell Can

supports. Average installation rates for the Propsetter and the

ACS are three times greater than rates achieved with

conventional wood cribbing. The installation rate with the

Burrell Can support depends on the logistics of unloading and

installation activities that require machinery to accomplish.

Installation rates vary from 30 to more than 50 supports per

shift. The Pumpable Crib that is poured in place currently

requires manual labor to lift the 55-lb grout bags from a pallet

positioned on a forklift near the pump. Heintzmann Corp. plans

to develop a batch system for the grout pumping activity, which
204

will all but eliminate the material handling efforts for this

support, with the exception of installing the forms to hold the

bag during pumping and dismantling them afterward. With a

three-person pumping crew underground and four people

handling material at the pump station aboveground, 50

Pumpable Cribs per shift have consistently been installed in a

test section at a mine site in western Pennsylvania. Even if

seven people were used on a crib construction crew, the number

of conventional four-point wood cribs constructed per shift

would probably be in the range of 35-40.

Table 7.—Transport parameters (normalized to 8-ft mining height)

Support system Support design Pieces per

support

Transport

volume,

ft3

/support

No. of

supports per

 supply car

Conventional wood cribs . . . . . four-point cribs1 . . . . . . . . . . . . 32 24.0 16

nine-point cribs1 . . . . . . . . . . . 48 36 11

Conventional concrete cribs . . Stopping block cribs (16 by 16

 in) . . . . . . . . . . . . . . . . . . . . .

32 16.2 24

SFR donut cribs . . . . . . . . . . . 32 29.9 15

SFR 24 -by 24-in solid crib . . . 72 32.0 12

SFR four-point crib (24×24 in) . 48 21.3 18

Hercules cribs . . . . . . . . . . . . . HM-4 . . . . . . . . . . . . . . . . . . . . 24 21 19

HM-9 and HM-9 (308) ...... 24 40 10

HM-16 . . . . . . . . . . . . . . . . . . . . 16 49.8 8

Link-N-Lock cribs . . . . . . . . . . . . 24-in Link-N-Lock . . . . . . . . . . . 64 18.7 21

27-in Link-N-Lock . . . . . . . . . . . 64 21.0 19

30-in Link-N-Lock . . . . . . . . . . . 64 23.3 17

36-in Link-N-Lock . . . . . . . . . . . 64 28.0 14

42-in Link-N-Lock . . . . . . . . . . . 64 32.7 12

48-in Link-N-Lock . . . . . . . . . . . 64 37.3 11

60-in Link-N-Lock . . . . . . . . . . . 64 46.7 8

Link-N-X cribs . . . . . . . . . . . . . . 24 in (standard design) . . . . . . . 32 9.3 42

27 in (standard design) . . . . . . . 32 10.5 37

24 in (high-capacity design) . . . 48 16.0 25

30 in (high-capacity design) . . . 48 20.0 20

36 in (high-capacity design) . . . 48 24.0 16

Lock-N-Load props . . . . . . . . . . Standard design . . . . . . . . . . . . 4 3.2 124

Propsetter . . . . . . . . . . . . . . . . . 8.5-in diameter . . . . . . . . . . . . . . 3 4.6 84

10.0-in diameter . . . . . . . . . . . . . 3 6.0 65

Star Props . . . . . . . . . . . . . . . . . 100 ton (12-in diameter) . . . . . . 3 9.2 43

60 ton (10-in diameter) . . . . . . . 3 6.7 58

Tri-Log cribs . . . . . . . . . . . . . . . . 30 in (standard design) . . . . . . . 48 19.8 20

36 in (standard design) . . . . . . . 48 23.7 17

30 in (high-capacity) . . . . . . . . . 48 29.7 13

36 in (high-capacity) . . . . . . . . . 48 35.6 11

48 in (high-capacity) . . . . . . . . . 48 47.5 8

Burrell Can . . . . . . . . . . . . . . . . . 18-in diameter . . . . . . . . . . . . . . 4 18.0 22

24-in diameter . . . . . . . . . . . . . . 4 31.6 12

30-in diameter . . . . . . . . . . . . . . 4 49.1 8

36-in diameter . . . . . . . . . . . . . . 4 70.4 6

Confined core crib (3-C) . . . . . . 36-in diameter . . . . . . . . . . . . . . 4 70.4 6

ACS or 55-ton prop . . . . . . . . . . Flat plate (8 in) . . . . . . . . . . . . . . 1 2.8 139

Timbers as header . . . . . . . . . . . 4 4.3 91

Pizza head plate . . . . . . . . . . . . 2 4.1 95

Pumpable Crib2 . . . . . . . . . . . . . 30-in diameter . . . . . . . . . . . . . . 1 3.9 100

Tekcrib . . . . . . . . . . . . . . . . . . . . 42-in diameter . . . . . . . . . . . . . . 1 7.7 51

Tekprop . . . . . . . . . . . . . . . . . . . 18-in diameter . . . . . . . . . . . . . . 1 12.8 31

Stretch Prop . . . . . . . . . . . . . . . . 6-ft collapsed length . . . . . . . . . 3 3.7 107

Yippi Prop . . . . . . . . . . . . . . . . . Standard design . . . . . . . . . . . . 1 2.8 140

1

Based on 6- by 6- by 36-in oak timbers.

2

Material assessment considers only that used underground, i.e., the bags for forming the support. It does not

include the hardware required to support the bags for filling since this hardware is used over and over again.

Since the supports were pumped from the surface, the grout material is not included in this assessment.
205

Table 8.—Support construction parameters (normalized to 8-ft construction height)

Support system Support design Weight per

piece, lb

Total

support

weight, lb

Construction

work, ft-lb

Installation

rate,2

 supports

per shift

Conventional wood cribs . . . . . . . . four-point cribs1 . . . . . . . . . . . . . . . . 40 1,280 5,760 15

nine-point cribs1 . . . . . . . . . . . . . . . . 40 1,920 8,640 12

Conventional concrete cribs . . . . . . Stopping block cribs (16 by 16 in) 52 1,664 8,055 19

SFR donut cribs . . . . . . . . . . . . . . . . 56 1,792 8,064 19

SFR 24- by 24-in solid crib . . . . . . . 53 3,816 17,172 9

SFR four-point crib (24×24 in) . . . . 53 2,544 11,448 13

Hercules cribs . . . . . . . . . . . . . . . . . HM-4 . . . . . . . . . . . . . . . . . . . . . . . . . 47 752 3,008 32

HM-9 (308) . . . . . . . . . . . . . . . . . . . . 44 1,056 4,752 26

HM-16 . . . . . . . . . . . . . . . . . . . . . . . 41 1,968 8,856 16

Link-N-Lock cribs . . . . . . . . . . . . . . . 24-in Link-N-Lock . . . . . . . . . . . . . . 13 813 3,658 24

27-in Link-N-Lock . . . . . . . . . . . . . . 15 941 4,234 24

30-in Link-N-Lock . . . . . . . . . . . . . . 17 1,069 4,810 19

36-in Link-N-Lock . . . . . . . . . . . . . . 21 1,318 5,933 16

42-in Link-N-Lock . . . . . . . . . . . . . . 25 1,574 7,085 16

48-in Link-N-Lock . . . . . . . . . . . . . . 29 1,824 8,208 15

60-in Link-N-Lock . . . . . . . . . . . . . . 37 2,330 10,483 14

Link-N-X cribs . . . . . . . . . . . . . . . . . 24 in (standard design . . . . . . . . . . 14 343 1,201 64

27 in (standard design) . . . . . . . . . . 16 518 2,333 45

24 in (high-capacity design) . . . . . . 13 638 2,554 32

30 in (high-capacity design) . . . . . . 18 854 3,417 30

36 in (high-capacity design) . . . . . . 22 1,152 4,282 21

Lock-N-Load props . . . . . . . . . . . . . Standard design . . . . . . . . . . . . . . . 44 88 513 96

Propsetter . . . . . . . . . . . . . . . . . . . . 8.5-in diameter . . . . . . . . . . . . . . . . . 132 168 820 48

10.0-in diameter . . . . . . . . . . . . . . . . 184 216 1,049 40

Star Props . . . . . . . . . . . . . . . . . . . . 100 ton (12-in diameter) . . . . . . . . . 247 301 1,668 40

60 ton (10-in diameter) . . . . . . . . . . 172 210 1,220 40

Tri-Log cribs . . . . . . . . . . . . . . . . . . . 30 in (standard design) . . . . . . . . . . 19 916 3,667 26

36 in (standard design) . . . . . . . . . . 24 1,133 4,531 21

30 in (high-capacity) . . . . . . . . . . . . 18 846 3,806 17

36 in (high-capacity) . . . . . . . . . . . . 22 1,593 7,167 14

48 in (high-capacity) . . . . . . . . . . . . 31 2,241 10,083 13

Burrell Can . . . . . . . . . . . . . . . . . . . . 18-in diameter . . . . . . . . . . . . . . . . . N/A 162 1,134 40

24-in diameter . . . . . . . . . . . . . . . . . N/A 270 1,890 40

30-in diameter . . . . . . . . . . . . . . . . . N/A 405 2,835 40

36-in diameter . . . . . . . . . . . . . . . . . N/A 567 3,969 40

Confined core crib (3-C) . . . . . . . . . 36-in diameter . . . . . . . . . . . . . . . . . N/A 567 3,969 40

ACS or 55-ton prop . . . . . . . . . . . . . Flat plate (8 in) . . . . . . . . . . . . . . . . . 151 151 680 64

Timbers as header . . . . . . . . . . . . . . 151 232 1,247 53

Pizza head plate . . . . . . . . . . . . . . . 140 178 896 53

Pumpable Crib . . . . . . . . . . . . . . . . . 30-in diameter . . . . . . . . . . . . . . . . . N/A 200 900 50

Tekcrib . . . . . . . . . . . . . . . . . . . . . . . 42-in diameter . . . . . . . . . . . . . . . . . N/A 160 720 60

Tekprop . . . . . . . . . . . . . . . . . . . . . . 18-in diameter . . . . . . . . . . . . . . . . . N/A 80 360 59

Stretch Prop . . . . . . . . . . . . . . . . . . . 6-ft collapsed length . . . . . . . . . . . . 50 77 231 69

Yippi Prop . . . . . . . . . . . . . . . . . . . . Standard design . . . . . . . . . . . . . . . 92 92 414 80

1

Based on 6- by 6- by 36-in oak timbers.

2

The installation rates may vary considerably due to the labor and equipment used in support construction and delivery of support

material to the working area. The numbers shown are representative installation rates. The installation rates are not normalized to

man-hours or effort. The support construction crew is generally two or three people, although the PumpableCrib installation currently

uses as many as seven people (all of whom work for the support manufacturer).

CONCLUSIONS

Material handling is an important aspect of secondary roof

support construction and more attention should be paid to it in

the support design and selection process. With more than

40,000 lost workdays attributed to timber-handling injuries in

the past 9 years, the construction of conventional wood cribs

and timber supports is the primary cause of injury to these

underground mine workers. Included in this paper is a detailed

summary of the various roof support systems and a description

of relevant material handling parameters to facilitate consideration of material handling factors in the selection of a

standing roof support system.

In recent years, several alternative support technologies have

been developed, which in addition to providing superior roof

control, offer material handling advantages. Surveillance data
206

show that the increase in use of these alternative support

technologies is consistent with the decreasing trend of material

handling injuries due to roof support construction in recent

years. The severity of these injuries has decreased by nearly

50% during 1995-98, which is when the use of alternative

support technologies attained proportions where they exceeded

the number of conventional cribs. Thus, these new support

technologies are having a positive impact on reducing material

handling injuries to coal miners.

Using the NIOSH lifting equation, which defines lifting

thresholds for various conditions and lifting scenarios, it is seen

that the weight of conventional crib timbers exceeds the recommended lifting threshold. Conversely, the weights of the engineered timber products is within the recommended lifting

thresholds and provides further confirmation that these lightweight materials are reducing material handling injuries. Systems that are installed with mechanical aid, such as the Burrell

Can Support (Burrell Mining Products) and the Star Prop

Propsetter (Strata Products USA), or pumpable support systems

that are installed in place in the mine, such as the Pumpable

Crib (Heintzmann Corp.), can substantially reduce the effort

required to install supports and thus dramatically reduce the risk

of injury to the mine worker. Therefore, depending on the other

parameters being held fairly constant, these alternative supports

should reduce the risk of musculoskeletal injuries due roof

support construction in underground mines.

When material handling is required, following some basic

lifting practices can make a difference in preventing injury.

Most importantly, the material should be held as close to the

body as possible to reduce the stress on the back, and twisting

of the body during the lifting process should be minimized and

avoided, if possible. Extra care must be exercised in the restricted environment of an underground mine to avoid injury.

Each miner is different, but everyone has a comfort zone in

being able to lift materials of a certain weight for a given

amount of time. The probability of injury increases when muscle fatigue occurs, so proper rest periods to avoid over exertion

can mean a lot in preventing injuries.

ACKNOWLEDGMENTS

The author acknowledges the efforts of Deno M. Pappas,

civil engineer, NIOSH Pittsburgh Research Laboratory (PRL),

who collected pertinent injury data pertaining to the construction of roof support systems in underground mines. These

data provide valuable insight into the type and magnitude of

injuries involved in support construction and were useful in

comparing injury trends for alternative support technologies

that have material handling advantages over conventional wood

and concrete cribs. The author also acknowledges Kim M.

Cornelius, industrial engineer; Lisa J. Steiner, safety engineer;

and Sean Gallagher, research physiologist, also with NIOSH

PRL, for their contributions in the application of the NIOSH

Revised Lifting Equation and for their research accomplishments pertaining to the biomechanics of lifting heavy materials.

Much of their work is being adopted for the roof support

construction application, which is addressed in this paper.

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coal mines: redesign of materials-handling tasks. Pittsburgh, PA: U.S.

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Waters TR, Putz-Anderson V, Garg A, Fine LJ [1993]. Revised NIOSH

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207

NIOSH SAFETY PERFORMANCE TESTING PROTOCOLS FOR STANDING

ROOF SUPPORTS AND LONGWALL SHIELDS

By Thomas M. Barczak1

ABSTRACT

The safety of mine workers depends on the proper installation of roof supports to prevent the ground from

collapsing into the working areas of an underground mine. As new support systems are developed, they need

to be properly evaluated to make sure that they are capable of providing adequate roof support before they are

first used in a mine. In addition to making certain that the supports meet basic safety criteria, the limitations

of the support need to be fully defined in order to avoid improper application of a particular support design.

The National Institute for Occupational Safety and Health (NIOSH) operates a world-class facility called the

Safety Structures Testing Laboratory. This laboratory contains a unique load frame, the Mine Roof Simulator,

which is capable of simulating the ground behavior in underground mines for conducting full-scale evaluations

of roof support systems. Safety performance testing protocols using the unique Mine Roof Simulator have

been developed for both standing roof support systems and longwall shield supports. The purpose of this paper

is to describe these test procedures. The protocol for standing roof supports incorporates seven test series:

( 1) uniform loading baseline tests, (2) height evaluations, (3) asymmetric loading, (4) biaxial loading, (5) load

rate studies, (6) active loading determination, and (7) static loading evaluations. For longwall shields, a fourseries test program that accurately simulates in-service conditions on a longwall face is proposed. This test

program consists of (1) transfer of horizontal load to the caving shield-lemniscate assembly (zero-friction test),

(2) point loading of shield joints due to lateral movement or rotation of the canopy, (3) evaluation of leg socket

and leg cylinder integrity, and (4) face-to-waste racking of the shield. In addition, an evaluation of the shield’s

hydraulic components will be conducted prior to the performance testing. These protocols will provide stateof-the-art safety performance evaluations of emerging support technologies, as well as a means to assess the

safety of both new and aging longwall shields. Hence, this effort will enhance the safety of mine workers by

ensuring that critical support elements are properly designed and that aging supports are retired before their

support capability is jeopardized.

1

Research physicist, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
208

INTRODUCTION

Ground control is one of the most fundamental aspects of

underground mining. Roof support systems are needed to

stabilize exposed mine openings and prevent collapse of the

mine roof. Without these critical support structures, the safety

of the miners would continuously be in jeopardy. Therefore, it

is essential that roof support systems be properly designed so

that they can provide adequate ground control in all

circumstances. 

The National Institute for Occupational Safety and Health

(NIOSH) is available to conduct safety performance testing of

emerging roof support technologies as they are developed to

assist manufacturers in meeting basic safety standards before

the supports are ever used in an underground coal mine. These

tests are also designed to ensure that the support is a viable roof

support system. Hence, in addition to evaluating basic safety

criteria, the safety performance tests are designed to determine

the limitations of the support system by evaluating the support

performance to failure under various loading conditions, so that

performance characteristics can be matched to ground behavior

in a particular mine in which the support system is installed.

These tests are conducted at the Safety Structures Testing

Laboratory in the unique Mine Roof Simulator load frame and

simulate actual in-service conditions in a mine. The tests are

conducted through cooperative agreements established with the

various support manufacturers. 

In the past 7 years, over 1,000 tests have been conducted on

various secondary roof support systems. As a result of this

effort,18 new support systems have been successfully adopted

by the mining industry, making a significant impact on longwall

tailgate support as alternatives to conventional wood and concrete cribbing. These include the following supports developed

by Strata Products USA: Hercules crib, Link-N-Lock crib,

Link-N-X crib, Propsetter support, Power Wedge, Rock Prop,

and Star Prop. Heintzmann Corp. developed the Alternative

Crib Support (ACS), the 55-Ton Prop, Quick Timber, and the

Pumpable Crib. Burrell Mining Products conducted tests on

The Can support. Fosroc Corp. developed the Tekcrib and

Tekprop supports. American Commercial, Inc., developed the

Tri-Log crib. Ferrocraft, Inc., developed the Stretch Prop and

other yieldable timber posts systems. Safety performance tests

were conducted on the YIPPI Prop (Western Support Systems)

and the Coal Post (Dywidag Systems International, Inc.). 

In addition to the development of innovative alternatives to

conventional wood and concrete cribbing, safety performance

testing protocols have been developed for longwall shield

supports. The unique loading capabilities provided by the Mine

Roof Simulator are especially suited to testing shield supports.

The caving mechanics of strata in longwall mining, and inservice loading conditions can be simulated much more realistically than is possible in a static load frame. Cyclic testing procedures have also been developed to evaluate the remaining life

of aging longwall shields.

The purpose of this paper is to describe the support testing

protocols developed for the unique loading capabilities of the

Mine Roof Simulator load frame, protocols that will improve

the safety of mine workers by helping design support systems

properly and by evaluating aging supports so they are not used

beyond their useful life span.

NIOSH SAFETY STRUCTURES TESTING LABORATORY

The Mine Roof Simulator is a servo-controlled hydraulic

press custom built by MTS Systems Corp. to U.S. Bureau of

Mines (USBM) specifications. It was designed specifically to

test longwall shields, and is the only active load frame in the

United States that can accommodate full-size shields.

A functional diagram of the load frame is shown in figure 1.

The load frame has several distinctive characteristics. The size

of the platens are 20 ft x 20 ft. The upper platen can be moved

up or down and hydraulically clamped into a fixed position on

the directional columns to establish a height for tests. With a

maximum vertical opening between the upper and lower platen

of 16 ft, the load frame can accommodate the largest shields

currently in use. Load application is provided by controlled

movement of the lower platen. The load frame is a biaxial

frame, capable of applying both vertical and horizontal loads.

Load actuators are equipped with special hydrostatic slip

bearings to permit simultaneous load and travel. This allows

vertical and horizontal loads to be applied simultaneously. The

capability to provide controlled loading simultaneously in two

orthogonal directions is unique at this scale.

Vertical load is applied by a set of four actuators, one on

each corner of the lower platen. Loads of up to 3 million

pounds can be applied in the vertical direction by upward

movement of the lower platen. Each actuator is capable of

applying the full 3 million pounds of force, so that the specimen

can be placed anywhere on the platen surface and the full

3 million capacity can be provided. The vertical (upward)

range of motion of the lower platen is 24 in.

Horizontal loading is applied by four actuators, with two

actuators located on both the left and right side of the load

frame just below floor level. These actuators act in pairs to

provide horizontal displacement of the lower platen in either a

positive or negative (x) direction. The horizontal range of

motion of the lower platen is 16 in.
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Figure 1.–Functional diagram of Mine Roof Simulator.

There is no programmable control of the lower platen in the

lateral horizontal axis (y-direction). The load frame has a

reactive capacity of 1.6 million pounds in this direction, but

loads can not be applied laterally. The range of motion of the

lower platen in this direction is ± 0.5 in.

The lower platen is controlled within six degrees of freedom

through the unstressed reference frame. This frame provides

feedback on platen displacements and rotations to the closedloop control system. Pitch, yaw, and roll of the lower platen are

controlled to keep the lower and upper platens parallel during

load application.

A shock absorber actuator is positioned on the left and right

sides of the lower platen. These shock absorbers will control

displacement of the lower platen to less than 0.1 in in the event

of a sudden failure of the support specimen. The shock

absorber action absorbs energy stored in the load frame so that

it is not unintentionally released to the test specimen.

Two hydraulic pumps provide up to 3,000 psi of pressure to

the vertical and horizontal actuators during load application.

The rate of movement of the lower platen is limited by the 

140-gal/min capacity of the hydraulic pumps. The maximum

platen velocity is 5 in/min, assuming simultaneous vertical and

horizontal displacement.

STANDING ROOF SUPPORT TESTING PROTOCOL

Standing roof supports are structures that are placed in a

mine entry between the roof and the floor. Their performance

can be described relative to three primary design factors:

(1) strength, (2) stiffness, and (3) stability. 

Strength – The strength of a roof support generally refers to its

ultimate load capacity. Hence, all supports are tested to failure

to determine the strength of the support.

Stiffness – Stiffness is a measure of how quickly a support

develops its load-carrying capacity and determined by

measuring support load capacity as a function of applied

convergence.

Stability – Stability is a measure of how long a support can

sustain its load-carrying capacity. The stability of a support

structure is affected by several parameters. These include (1)

aspect ratio of the support, (2) boundary conditions established

with the load frame at the roof and floor contact, (3) direction

of load application, (4) quality and properties of the specimen,

and (5) rate of loading. 

STANDARDIZED TESTS FOR STANDING SUPPORTS

TEST SERIES I – UNIFORM LOADING

BASELINE TESTS

Objective – Establish baseline performance of a support under

ideal loading conditions.

Test Requirement – Simulate roof-to-floor convergence by

applying uniform loading to the support element. The response

of the support structure is measured relative to its stiffness,

strength, and stability.

Test Procedure – A representative support is placed in the

Mine Roof Simulator with full roof and floor contact to

establish uniform loading on the support. A controlled vertical

displacement at a rate of 0.5 in/min is applied to the support

system by the load frame to simulate convergence of the mine

roof and floor. The applied load is measured as a function of

vertical displacement to determine the stiffness of the support.

Convergence continues until the support (1) becomes unstable,

(2) sheds load to the point where the support provided is

inadequate, or (3) until the full 24-in stroke of the load frame is
210

Figure 2.–Asymmetric loading configurations.

reached. Ultimate strength and complete performance profiles

are determined by plotting support load versus applied

displacement. 

TEST SERIES II - IMPACT OF SUPPORT SIZE

ON STABILITY AND CAPACITY

Objective B Determine the impact of the size of the support on

its capacity and define proper support sizes that will ensure

stability through a useful convergence. 

Test Requirements B Vary support sizes and provide uniform

loading through controlled roof-to-floor convergence. The capacity of the support as a function of the support area will be

determined from this suite of tests. The stability of some support systems is largely governed by the aspect ratio or the

height-to-width ratio of the support. When this is a design

parameter, the support will be evaluated at several heights

representing various aspect ratios to determine the limits of the

support is stability. Standard heights are 4, 6, 8, 10, and 12 ft.

 Typically, the support is widened to maintain stability at higher

operating heights. The goal of the test is to determine an acceptable aspect ratio range over which the support will maintain

stability at all recommended operating heights. For example,

tests on conventional wood crib supports have determined that

the aspect ratio should be maintained between 2.5 and 5.0, with

4.3 considered an optimum for uniform load conditions.

Test Procedure B The test procedure is basically the same as

in the first test series. A representative support is placed in the

Mine Roof Simulator with full roof and floor contact to establish uniform load on the support. A controlled vertical displacement is applied to the support system to simulate convergence. Convergence continues until the support (1) becomes

unstable, (2) sheds load to the point where the support provided

is inadequate, or (3) until the full 24-in stroke of the load frame

is reached. The ultimate strength and capability of the support

needed to sustain load resistance while yielding will be determined by analysis of the load-displacement profile. 

TEST SERIES III - ASYMMETRIC LOADING

Objective – Determine the impact of asymmetric loading on the

stability and overall support capability of the support. 

Test Requirements – Simulate asymmetric loading conditions

that occur with uneven roof and floor contact or because of

wedging the support in place. Figure 2 illustrates four asymmetric

loading configurations that can be applied to standing roof

supports. Figure 3 shows some examples of actual supports being subjected to these asymmetric loading conditions.

Another condition that creates asymmetric loading is floor

heave. Floor heave is simulated by creating a foundation that

rotates as support load is developed. This is accomplished by
211

Figure 3.–Examples of asymmetric loading conditions.

placing the roof support structure on a rigid steel plate that is supported on one side by a soft (crushable) support and a stiff (rigid)

support on the opposite side. Figure 4 illustrates this arrangement

and an example of a test conducted in the Mine Roof Simulator.

Test Procedure – A support is placed in the load frame. A

specific roof and floor contact is established in accordance with

the diagrams shown above by strategically placing contact blocks

at the roof and floor interface. A controlled vertical displacement

is applied to the support. Convergence continues until the support

(1) becomes unstable, (2) sheds load to the point where inadequate support is provided, or (3) until the full 24-in stroke of

the load frame is reached. The ultimate strength and capability

of the support to sustain load resistance while yielding will be

determined by analysis of the load-displacement profile. Upon

completion of this test, another support is installed and another

contact configuration is established to evaluate a different

asymmetric loading condition. The test procedure is then repeated for this and any other asymmetric loading configuration.

TEST SERIES IV - BIAXIAL LOADING

Objective – Determine the impact of horizontal loading on

support capability. 

Test Requirements – Simulate both vertical (roof-to-floor)

convergence as well as lateral movements associated with

bending or buckling of laminated roof or floor structures as a

result of horizontal stress. This is accomplished by moving the

floor of the load frame simultaneously in both vertical and

horizontal directions creating a load vector in which the base of

the support is moved laterally with respect to the top of the roof

support at the same time the support is being squeezed by roofto-floor convergence (see figure 5).
212

Figure 4 - Floor heave simulation and examples of support testing.

Figure 5.–Biaxial load conditions and example of support being subjected to biaxial loading.

Test Procedure – A support is placed in the load frame.

Typically, full roof and floor contact is utilized for this test

series. The Mine Roof Simulator is commanded to apply a ratio

of vertical to horizontal displacement. The standard ratio (vertical to horizontal displacement) is 3:1, although the ratio can be

varied if desired. The applied biaxial convergence continues

until the support (1) becomes unstable, (2) sheds load to the

point where inadequate support is provided, or (3) until the full

24-in vertical stroke of the load frame is reached. The ultimate

strength and capability of the support to sustain load resistance

while yielding will be determined by analysis of the loaddisplacement profile. If the support stability is sensitive to

changes in the aspect ratio as determined in test series II, then

the support height will also be varied.

TEST SERIES V - LOAD RATE STUDIES

Objective – Some supports have a tendency to provide greater

load resistance as the loading rate is increased. The objective

of this test series is to determine the impact of loading rate on

the support’s behavior.

Test Requirements – Vary loading rate by controlling the

applied roof-to-floor convergence. The Mine Roof Simulator

can control the rate of roof-to-floor from 0.1 to 5.0 in/min.

Test Procedure – Baseline test data were established in test

series I at the standard loading rate of 0.5 in/min. To establish
213

Figure 6.–Violent failure of concrete crib at 5 in/min applied

convergence.

a load rate profile, supports are tested to failure at least two

additional rates. Typically, rates of 0.1 and 5.0 in/min are utilized. Support load as a function of convergence is then compared for the different loading rates, and the impact of loading

rate on the stability of the support and the nature of the failure

are documented. Figure 6 illustrates a concrete crib exploding

during a high rate of loading. 

TEST SERIES VI -ACTIVE LOADING DETERMINATION

Objective – Some roof supports are capable of providing an

active roof load during installation of the support. The objective of this test series is to determine the active loading

capability of those supports.

Test Requirements – Measure the active roof loading generated by a support during its installation. 

Test Procedure – A load cell is placed on top of the support to

obtain a more accurate measure of applied roof loading,

particularly when the measured active roof loads are expected

to be less than 20 kips. The load frame platens remain stationary during the test. An effort is also made to determine

whether active loading remains constant or is shed over time

once the support is installed. Hence, a plot of active roof

loading as a function of time is made, and a decay rate is determined. The amount of time can depend on type of support,

but the initial period is 30 min.

TEST SERIES VII - STATIC LOADING

EVALUATIONS

Objective – Static loads are used to assess creep and relaxation

in support material construction. The objective of this test series

is to determine the creep and relaxation properties of a support.

Test Requirements – Creep is the continuation of deformation

after a static load has been applied. To measure creep, a

constant force must be applied to the support. Relaxation is the

opposite of creep. Relaxation is the reduction in stress or load

after an applied displacement. Hence, the test requirement to

measure creep is maintaining constant displacement. 

Test Procedure – For the creep study, a support is placed in the

load frame. The load frame is operated in force-control, and a

load is applied and held constant for an extended period. The

change in displacement is then measured to determine the rate

of creep.

For the relaxation study, the load frame is operated in displacement control, and the support is loaded through a designated convergence, that is held constant by the load frame. The

change in support load is then measured as a function of time to

determine the relaxation properties of the support.

NIOSH SAFETY PERFORMANCE TESTING PROTOCOL FOR LONGWALL SHIELDS

NIOSH also conducted shield performance tests in the

Safety Structures Testing facility (figure 7). The Mine Roof

Simulator can simulate in-service loading conditions on shields

more accurately than static load frames. It is the only active

load frame in the United States with sufficient size and load

capacity to accommodate shield testing and allows realistic and

cost-effective shield evaluations by combining both vertical and

horizontal (racking) loads into a single load cycle. 

The ultimate goal of the NIOSH shield testing program is to

ensure the safety of mine workers by ensuring that new shields

are adequately designed and that aging or damaged shields

retain adequate structural integrity for continued use. This

section describes the shield testing protocols, developed through

extensive studies of shield mechanics and performance tests.

SIMULATION OF IN-MINE SERVICE CONDITIONS

There are two basic aspects to shield loading. The initial load

condition is determined by actively setting the shield against the

mine roof and floor. Subsequent loading is produced by the movement of the surrounding strata during the caving process and the

associated internal forces developed within the support structure.
214

Figure 7.–NIOSH Mine Roof Simulator.

Figure 8.–Horizontal movement of canopy toward longwall

face as the shield is set against the mine roof.

As the shield is set against the mine roof and floor, there is

a tendency for the canopy to be displaced horizontally relative

to the base (figure 8). This is due to the resultant horizontal

component of the leg forces, which causes either slippage of the

canopy along the roof interface or displacement (compaction)

of fractured strata or debris immediately above or below the

shield. The Mine Roof Simulator accurately simulates this

behavior by allowing the floor of the load frame to move horizontally and transfer horizontal load from the horizontal

component of the leg forces to the caving shield-lemniscate

assembly. When a shield is tested against a rigid frame, the

canopy and base are restrained from moving horizontally. This

restraint eliminates load development in the caving shieldlemniscate assembly and therefore does not properly simulate

in-mine service conditions.

Once the shield is set against the mine roof and floor, load

development within the shield is controlled by—

(1) Contact configuration established with the mine roof

and floor,

(2) Vertical displacement of the canopy relative to the

base induced by deflection of the main roof beam and

the weight of the fractured immediate roof strata being

supported by the shield (figure 9), 

(3) Face-to-waste movement of the immediate roof as the

strata break into disjointed blocks because of face abutment loading and loss of confinement (figure 10), and

(4) Waste-to-face loading induced by gob material acting

on the caving shield and/or the internal forces developed within the shield resulting from leg forces and

component reactions (figure 11) and lateral loading due

to skewing of the canopy caused by setting against

adjacent shields, inclination of the face, or rotation of

the canopy due to uneven roof and/or floor conditions

(figure 12).

 
215

Figure 9.–Vertical shield loading induced by deflection of

main roof and weight of damaged immediate roof.

Figure 10.–Horizontal shield loading induced by face-towaste movement of immediate roof.

Figure 11.–Horizontal loading toward coal face induced by

gob loading on caving shield.

Figure 12.–Lateral loading induced by uneven roof contact.

NIOSH STANDARDIZED SHIELD TEST PROCEDURES

Standard tests consist of an evaluation of hydraulic

components under static loading conditions followed by a series

of cyclic tests to evaluate the structural integrity of the shield.

For each test, the shield is positioned in the load frame in an

orientation consistent with the objectives of the test. Prior to

cyclic loading, the shield is actively set against the load frame

platens by pressurization of the leg cylinders to some nominal

load, typically 50 to 75 bar. Cyclic loading is provided by

controlled displacement of the Mine Roof Simulator load frame

floor against a stationary roof. Each load cycle consists of

ramping the load, a hold, an unloading ramp, and a hold. A

combination of vertical and horizontal displacements are

applied, often simultaneously, to produce the required load

conditions. The loading rate is dependent on the shield stiffness

and capabilities of the load frame. Two load cycles per minute

are a typical loading rate. The loading profile in each test series

is designed to maximize total shield loading. Hence, load

profiles are typically chosen that provide load equal to the yield

load rating for the shield. Since the Mine Roof Simulator is an

active load frame, there is no need to exceed the rated capacity

of the shield to account for friction effects within the support

structure. A minimum of 5,000 cycles for each test series is

recommended, but this number may be varied depending on the

customer's needs.

HYDRAULIC COMPONENT EVALUATIONS

A series of tests are conducted to determine the performance

and condition of the leg cylinders and shield hydraulics.

Yield setting – The shield is loaded until each yield valve on

all leg cylinders opens. The recorded pressure and maximum

shield rating for the designated operating height are determined.

Leakage Test –The leg cylinders are pressurized to some

nominal load and held for 30 min. During the hold, the

pressures are monitored. Leakages are evaluated to determine

the source of the leakage.
216

Figure 13.–Test to determine defective staging valve.

Figure 14.–Applied loading for test series I.

Figure 15.–Shield response to test series I loading.

Staging Valve Test – The bottom stage on all leg cylinders is

fully extended, and the shield is actively set against the load

frame platens. Additional load is then applied to the shield

while hydraulic pressure is monitored. If the staging valve is

working properly, the pressure in the bottom stage should not

increase until the force in the upper stage equals the setting

force developed in the bottom stage (figure 13).

TEST SERIES I – TRANSFER OF HORIZONTAL LOAD

 TO THE CAVING SHIELD-LEMNISCATE ASSEMBLY

Objective – Minimize external horizontal load acting on the

shield to ensure that horizontal components of the leg forces are

transferred to the lemniscate links, thereby maximizing load

development in the caving shield-lemniscate assembly.

Test Requirements – The canopy must be free to displace

horizontally with respect to the base to allow the caving shieldlemniscate assembly to participate to a degree consistent with

underground shield behavior. This is accomplished by commanding the floor of the load frame to move horizontally with

respect to the roof in a direction and magnitude consistent with

the resultant leg force. Main roof loading and deflection of the

immediate roof beam are simulated by controlled vertical

displacements. The applied displacement and associated shield

response are shown in figures 14 and 15.

Canopy and Base Contacts – A four-point contact on the

corners of the canopy is used to maximize bending produced by

the increase in leg pressure. A three-point canopy contact

where one of the rear contacts is removed can also be used to

further intensify stress development in the canopy. Base contacts are located at the ends of each base section to maximize

bending in the base. 
217

Figure 16.–Lateral loading caused tilting of leg cylinders

and lemniscate links during test series II.

Figure 17.–Illustration of point loading in lemniscate link

joints due to tilting of lemniscate link in test series II.

Test Procedure

1. The shield is set against the load frame roof and floor at 50-

to 75-bar leg pressure using an external hydraulic power

supply.

2. The floor of the load frame is moved horizontally in a direction that eliminates the horizontal load applied by the load

frame during the setting operation, which causes this horizontal load to be transferred to the shield components. The

elimination of the external horizontal restraint moves the resultant force acting on the base forward, which intensifies

toe loading, a critical load condition for two-leg shields. 

3. Cyclic loading is initiated by a controlled vertical and

horizontal movement of the lower platen of the load frame,

inducing a combined vertical and waste-to-face displacement

of the canopy relative to the base. The horizontal platen

movement is calibrated to minimize horizontal load restraint

provided by the load frame throughout the loading cycle.

For two-leg shields, this requires the canopy to be displaced

in a faceward direction at a rate that is proportional to the

increasing horizontal component of the leg force developed

from vertical closure. The result of these actions is that the

caving shield-lemniscate assembly is fully loaded to provide

internal equilibrium within the shield.

TEST SERIES II – POINT LOADING OF SHIELD

JOINTS

DUE TO LATERAL MOVEMENT OR ROTATION

OF CANOPY 

Objective – To maximize loading in the various shield joints

and component clevises by causing point load conditions due to

tilting of the lemniscate links caused by lateral movement of the

canopy relative to the base.

Test Requirements – Joint wear is the most common problem

causing premature shield retirement. The requirement for this

test is to induce a resultant load vector that skews the canopy

laterally with respect to the base (see figure 16). The shield

joints have a single degree of rotation, much like a person's

knee functions. Stress on the connecting pins is intensified

when the canopy is skewed laterally, causing partial contact of

the connecting pins within the clevis (figure 17). 

Canopy and Base Contacts – A three-point canopy contact

with one contact omitted from the rear canopy corner is used to

further maximize twisting of the canopy and connecting joints

(figure 18A). The outside base, away from the direction of the

tilt is supported at the ends, while the inside base in the direction of the tilt has full contact (figure 18B). 

Test Procedure

1. The shield is positioned in the load frame so that the direction of applied horizontal displacement (loading) is across

the canopy, as shown in figure 16. 

2. The shield is set against the load frame roof and floor at 50

to 75 bar of leg pressure.

3. The canopy is displaced laterally with respect to the base,

causing the leg cylinders and lemniscate links to tilt toward

the direction of applied lateral loading.

4. Cyclic loading is initiated by the active load frame applying vertical displacement while maintaining the lateral

displacement of the canopy with respect to the base. If

necessary, the internal load can be increased during the

vertical load application.
218

Figure 18.–Setup for test series II. A, Canopy contact; B, base contact.

Figure 19.–Setup for test series III. A, Base contact B, canopy contact.

TEST SERIES III – EVALUATION OF LEG SOCKET

AND LEG CYLINDER INTEGRITY

Objective – Maximize stress development in the leg socket

welds and expose the leg cylinder seal and piston to maximum

side loading at full-stage extension.

Test Requirements – Failure of the leg sockets is a common

shield problem. The leg socket is a casting that is welded along the

top four sides to the side rib plates of the base and to horizontal

stiffening plates in the base construction. The canopy construction

is similar, except that unlike the base, the canopy is a single unit

with additional stiffening plates built into the structure. The test

requirement is to induce maximum loading into the welds. 

Canopy and Base Contacts – For the base structure, plates are

cut so that their width is approximately 50 mm less than the

inside dimension of the side rib plates (figure 19A). This contact arrangement requires that the full load be supported by the

bottom plate without being carried directly through the side rib

plates. Therefore, the plates are designed to maximize stress

development in the rib socket welds. The base contact plates

are spaced 1 m apart during loading to simulate steps in the

floor caused by shearer cuts. A centerline canopy contact is

established as shown in figure 19B. The contact is positioned

between the leg sockets to induce transverse bending of the

canopy structure and focus loading on the leg socket welds.

Test Procedure

1. The shield is configured so that the upper stage of the leg

cylinders are at full extension (figure 20). The transverse

bending of the canopy with full extension of the upper leg

cylinder staging will cause maximum side loading of the

piston and seals.

2. The shield will be set against the load frame with 50 to 75

bar of leg pressure.

3. Cyclic loads are applied by controlled vertical displacement of the load frame platens. Horizontal positioning of

the canopy and base will be restrained by the load frame

during load application. The shield is cycled between the

setting load and yield load.
219

Figure 20.–Upper stage of leg cylinders is

fully extended during test series III.

Figure 21.–Canopy is displaced toward the gob in test

series IV to cause face-to-waste racking of shield.

TEST SERIES IV – FACE-TO-WASTE RACKING 

OF SHIELD (OPTIONAL)

Objective B Simulate the effects of roof strata pushing toward

the gob.

Test Requirements – Induce a load vector that produces faceto-waste racking of the canopy with respect to the base. The

caving shield-lemniscate assembly is designed to alleviate

bending moments on the hydraulic leg cylinders by absorbing

all horizontal loads acting on the shield. This condition is

unlikely to occur in two-leg shields except at high operating

heights where the leg and lemniscate link orientation is closer

to vertical. Hence, this test requirement is considered optional

depending on shield design and kinematics of the shield.

Canopy and Base Contacts – Full canopy and base contact is

utilized to facilitate frictional contact along the canopy and base.

Test Procedure

1. The shield is positioned in the load frame so that the

direction of applied horizontal displacement of the canopy

is toward the gob (figure 21).

2. The shield is set against the load frame roof and floor at

approximately 50 to 75 bar of leg pressure.

3. Cyclic loading is initiated by the active load frame applying a combined face-to-waste and vertical displacement of

the canopy relative to the base. Vertical displacement is

applied to the degree necessary to sustain horizontal

loading. Horizontal displ acement is applied until the legs

reach yield load. Since the horizontal displacement is in

the opposite direction to that in test series I, the lemniscate

link force is also opposite, with the front link acting in

tension and the rear link acting in compression. This

change in state of stress produces maximum wear and

fatigue loading. 

SHIELD INSTRUMENTATION AND DATA ACQUISITION

To monitor load development through each of the shield

components, strain gages are installed at selected locations on

the various shield components, as described below. The primary purpose of the strain gages is to monitor load transfer

through each of the shield components. The gages are not necessarily installed in areas where stress concentration is greatest.

Signal conditioning and data acquisition are provided by a data

acquisition system. Typically, each sensor is sampled once a

second to provide a reasonably complete load profile. Sampling

rates up to 10 kHz are available if needed to assess failure

developments. Both historical and real-time observations of the

data are possible through the data acquisition system. The

shield is inspected after every 1,000 load cycles for structural

damage. A dye penetrant can be used to assess crack developments. The utilization of magnetic flux, x-ray, or ultrasonic

technologies is beyond the scope of the standard test program

and will require additional funds. 

Canopy – Two to four gages are installed on the main vertical

ribs of the canopy structure to assess bending strains. One gage
220

SHIELD SAFETY PERFORMANCE TESTING 

FINAL REPORT

1. Scope of Work

2. Testing Objectives and Simulation Methods

Objectives

Simulation of the In-Mine Service Conditions

Standardized Shield Tests and Exceptions

3. Shield Instrumentation and Data Acquisition

4. Test Results

Hydraulic Component Evaluation

Cyclic Tests

Test Series I

Test Series II

Test Series III

Test Series IV

4. Failure Assessment and Problem Report

5. Conclusions and Recommendations

is typically installed near the leg connection where the bending

moment is the largest. The other gage is typically installed forward of the leg connection. Strain gages are generally installed

on both the left and right side of the canopy. 

Caving shield – A strain gage is installed near the canopy

clevises and/or the lemniscate link clevis on the main loadtransferring members of the caving shield.

Lemniscate links – One or two strain gages are installed on

each of the lemnscate links to measure load development in the

lemniscate links. Both axial and bending-induced strains are

measured. Gages on both the top and bottom surface are used

in link designs that promote bending of the link structure. 

Base – A series of two or three strain gages are installed on the

inside and outside ribs of each base fabrication to measure

bending in the base sections. Gages are also applied to the

bridge connecting the two base sections in split-base designs. 

Leg Sockets – Gages are installed on the plate sections that

support the leg socket in both the canopy and base. 

Hydraulic – Pressure transducer is installed in each hydraulic

cylinder to measure pressure development during loading.

Cycle count – The number of cycles is counted automatically

by tracking leg pressure development. 

DATA PRESENTATION

The applied loading and strain developments are monitored

during each load cycle. A full profile of strain development

during the loading cycle will be recorded at 100-cycle increments. Ten strain profiles collected during 1,000 loading cycles

will be plotted on a single graph for comparison. Maximum

and minimum strains will be recorded for each loading cycle.

 The maximum and minimum strain values will be plotted in

groups of 5,000 loading cycles to examine trends over a sustained loading period. NIOSH reserves the right to modify this

standard data presentation plan when extenuating circumstances

dictate that other data presentations would be adequate or more

appropriate.

SHIELD PERFORMANCE TESTING REPORT

A performance test report is provided as shown below. The report describes how the tests were conducted and the results of

the tests. A failure assessment is made documenting time of the failure and the component(s) involved.
221

CONCLUSIONS

The safety of mine workers depends heavily on roof support systems that prevent the unintentional collapse of ground

in both working and access areas of the mine. Support manufacturers continually strive to develop new support technologies

that provide more effective roof support at less cost and with

less effort to install. It is imperative that these prototype support technologies be thoroughly evaluated to make sure that

they meet required design criteria for use in various underground mine conditions.

The availability of the NIOSH Safety Structures Testing

Laboratory with the unique Mine Roof Simulator load frame

provides support manufacturers and mine operators with the

most precise simulation of underground conditions to ensure the

safety of their products prior to installing prototype systems.

The safety performance testing protocols developed by NIOSH

for application in this world-class facility are based on years of

research into support design and testing requirements. As such,

they are believed to provide the best possible evaluation of

support technology. 

In recent years, numerous roof support technologies have

been successfully developed and evaluated at the NIOSH Safety

Structures Testing Laboratory utilizing the protocols described

in this paper. In the past 7 years, over 1,000 tests have been

conducted on various secondary roof support systems. As a result of this effort,18 new support systems have been successfully introduced to the mining industry, making a significant

impact on longwall tailgate support as alternatives to conventional wood and concrete cribbing.

The NIOSH Safety Structures Testing Laboratory provides

an opportunity for coal operators andsupport manufacturers to

have shields tested domestically. The Mine Roof Simulator is

unique in its capabilities to apply active vertical and horizontal

loads simultaneously, providing realistic simulations of underground load conditions. Since the Mine Roof Simulator more

accurately simulates in-mine service conditions than static

frames, testing at the Safety Structures Testing Laboratory reduces the risk of premature failures due to poor design. In

addition to performance-testing new shields, the remaining life

of aging shields can be determined with more confidence and

can provide an engineering basis for shield retirement and new

shield procurement.

In summary, the benefits of shield performance testing

using the unique capabilities of the Safety Structures Testing

Laboratory includeC

! Unbiased assessment of support performance.

! A location in the United States that improves access to

mine operators.

! Freedom of control over the test program.

! NIOSH knowledge of shield mechanics and design issues.

! Capability for active loading as opposed to static-frame

testing.

! Controlled loading that simulates in-mine service conditions accurately.

! Participation in research programs to improve shield

design and operation.

Further information concerning utilization of the Safety

Structures Testing Laboratory can be obtained by contacting

Tom Barczak at (412) 386-6557, by fax at (412)386-6891, or

by email at thb0@CDC.gov.
223

EXAMINING LONGWALL SHIELD FAILURES FROM AN ENGINEERING

DESIGN AND OPERATIONAL PERSPECTIVE

By Thomas M. Barczak1

ABSTRACT

Longwall operators are again pushing the envelope in terms of life expectancy for longwall shields. Stateof-the-art shields are now expected to last more than 60,000 loading cycles, twice the life expectancy compared

with those of a decade ago. A review of trends in shield design shows that shields continue to increase in both

size and capacity. Some state-of-the-art shields now weigh over 30 tons and provide up to 1,200 tons of

support capacity. Although life expectancy has increased and modern shields are structurally more reliable,

premature failures do still occur. This paper provides an engineering and operational assessment of shield

design and provides key points to observe in what causes premature shield failures. Design practices to

improve structural margins of safety that will prevent premature failures from occurring are also examined.

A survey of recent shield failures is provided, as well as trends in shield design and how they might impact

the performance and longevity of a shield. Hydraulic failures are more common than structural failures.

Although hydraulic failures occur on all aging longwall shields, they often go undetected for long periods,

resulting in degraded support capacity that can lead to serious ground control problems. The fundamentals

of shield hydraulics are described in order to evaluate hydraulic failures that plague all shields at some point

in their service life, and practical methods to detect hydraulic failures are examined. The paper concludes with

recommendations for inspecting damaged shields and safety precautions regarding their continued use.

1

Research physicist, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
224

INTRODUCTION

The decline in the number of longwalls from a high of 118

in 1985 to 62 in 1999 has forced a reduction in the number of

shield manufacturers through mergers from 8 to 2. The two

major suppliers of shields to the U.S. market are Joy and DBT

America (formerly Mine Technik America or MTA). Their

share of the market is split fairly evenly: Joy has 32 faces and

DBT America has 30 faces [Fiscor 1999]. The author believes

that the merger of the shield manufacturers has been beneficial

to shield design in two ways. First, the merger has allowed the

best of the U.K. shield designs (Joy) and the best of the German

designs (DBT America) to be brought forward. By combining

the design strengths from each company, the result has been an

improvement in both structural design and control system

technology. Second, the fierce competition during the decade

of the 1980s may have forced manufacturers to "cut corners"

relative to design in order to compete for sales, resulting in

deficiencies in design that led to premature structural failures

and less than adequate control technology. Thus, while competition is generally healthy, resulting in lower prices, there

needed to be a better balance of quality and price than existed

in the 1980s. The hard reality is that shield manufacturers have

to make a profit in order to produce a quality product and to

stand behind this product with warranties; this capability was

jeopardized in the past and has been improved with the recent

mergers.

A survey of the longwall industry was conducted to evaluate

problems with current shield technology and future needs. The

survey indicates that shields are lasting longer than ever before,

but that premature failures still occasionally occur and some

mistakes from the past continue to be made in recent shield

designs. Structural failures tend to be the most catastrophic,

often requiring modifications to the original shield design.

These failures can cause considerable downtime and loss of

production, and expose the mine workers who must change out

the damaged components to increased risk of injury. Hydraulic

failures are common to all aging supports. These failures often

go undetected, resulting in degraded support performance that

can lead to serious ground control problems. Methods to detect

these hydraulic failures are discussed later in this paper.

Due to the maturity of shield technology and the always

present economic pressures of the mining industry, longwall

operators are now keeping shields in service longer than ever

before. In addition, due to the increases in longwall productivity, the number of operating cycles per year continues to

increase. The result is that the operators are again pushing the

envelope in terms of extending the life expectancy of the

supports. Today's shields are expected to last 60,000-70,000

cycles; 10 years ago the life expectancy was about 35,000

cycles. Thus, while fatigue failures were not a design issue in

the early generation of shield supports, they have become

critical to the survival of most longwall operators in an ever

increasingly competitive market. Much is being learned about

the behavior of aging shields, which are kept in service long

enough to fail from fatigue as opposed to failure from poor

design or replacement before the end of their useful life due to

technological improvements.

Deciding when to retire an aging longwall face and the

specifications for a new shield can be a critical decision for any

longwall operator. Most mines do not have structural engineers

that can actively participate in these decisions. An overview of

the fundamental engineering aspects of shield design is

provided in the paper, as well as key points that should be

considered in the design process to avoid premature structural

failures. A primary goal of this paper is to provide mine operators with a better sense of design issues and engineering

mechanics that are relevant to shield design. The intent is not

necessarily to have the operators learn enough to know all the

answers, but to provide them with enough insight so they know

what questions to ask and what to look for relative to failures

that may occur. These insights into shield design should also

be beneficial to the Mine Safety and Health Administration in

evaluating the safety of longwall shields.

The paper concludes with some practical recommendations

regarding what to look for and actions to take when failures do

occur. These key points will help mine operators examine failures and provide responsible actions to ensure the safety of the

mine workers when shield performance is degraded.

RECENT TRENDS IN SHIELD DESIGN

The basic shield structure has remained unchanged for the

past 20 years (figure 1), although the structures have grown dramatically in size and capacity. There have also been technological improvements in electrohydraulic control systems that have

dramatically impacted the operation of the support. Consequences of these changes in shield design are discussed below.

MATERIAL SPECIFICATIONS

The German shield companies (currently conglomerated

under DBT) have in recent years promoted the use of highstrength steels (>100,000-psi yield) to minimize component

cross-sectional dimensions. This trend continues in the present
225

Figure 1.—Diagram of shield components.

Figure 2.—Trend showing increase in support capacity.

Figure 3.—Distribution of capacity for shields operating in 1999.

with both DBT and Joy Technologies. High-strength steel

applications were first used in canopy structures to minimize

the cross-sectional thickness of the canopy in low- to moderateseam applications. Today, high-strength steel in all shield designs are necessary to provide the required strength for various

shield components in high-capacity designs. Although the

added strength has helped to extend shield life, the highstrength fabrications are more susceptible to brittle failure and

catastrophic fatigue failures. Another consequence of some

high-strength steel applications is that special welding practices

are required (heat control, etc.), which makes underground

repairs more difficult.

CAPACITY

Support capacity has continued to increase throughout the

history of longwall mining. This trend for the past 15 years is

shown in figure 2 [Fiscor 1999]. Average support capacities in

the United States have increased by nearly 50% since 1985 to

an average support capacity at yield of 768 tons for the 62

operating longwalls in the United States in 1999. Twelve installations (19%) of the current longwalls employ shields with

capacities greater than 900 tons, and 19 installations (50%)

have capacities between 700 and 900 tons. The current

distribution of shield capacities is shown in figure 3. The

highest capacity shield used in the United States is 1,170 tons

[Fiscor 1999]. As shown in figure 2, maximum shield

capacities have evolved from 800-ton shields, which were

common from 1985 to 1990, to 1,200-ton shields in 1999.

The increase in capacity has significantly impacted shield

design and ground control capability. In order to provide the

increased support capacity, larger diameter leg cylinders had to

be utilized. One consequence of the larger diameter leg

cylinders has been a proportional increase in shield stiffness.

This increase means that the load developed in the support

structure will occur at less displacement or face convergence.

Although the increased stiffness may provide superior control

of the immediate roof, main roof weighting that causes

irresistible (in terms of the shield capacity) convergence on the

longwall face will result in greater shield loading with the

stiffer shield design. As a result, it is not uncommon for the

high-capacity shields to be fully loaded to yield capacity just as

often as the lower capacity designs that they replaced. An

example is illustrated in figure 4, where a 500-ton, 800-ton, and

1,000-ton shield are all fully loaded when 0.5 in of convergence

occurs.

Another consequence of the larger diameter leg cylinders is

that the leg socket must be designed to accept greater loads.

Historically, the leg socket has been a source for premature

shield failures, and the increased loading makes the design even

more demanding. In addition, the larger diameter leg cylinders

require a wider and longer socket, which places further

demands on the design of the socket and load transfer to the

base and canopy structures. In general, the larger diameter

cylinders have resulted in wider support components, which are

more susceptible to torsion loading than before.

Several models have been developed over the years to

determine support capacity requirements. While these concepts

attempt to capture the support and strata interaction principles,

state-of-the-art shield capacities cannot be justified based on

these concepts. A "bigger is better" attitude has prevailed,

which has been promoted by the manufacturers largely because

of the demands of mine operators to improve the life

expectancy of the shields. Thus, it is the life expectancy issue,

along with ground control requirements, that have controlled

recent developments in longwall shield design and capacity

determinations.
226

Figure 4.—Setting load and load developed due to 0.5 in of

roof movement for shields of various capacities.

SHIELD TYPE

There has been a steady increase in the use of two-leg

shields in favor of four-leg shields during the past decade, and

two-leg shields are becoming the favored support worldwide.

In 1997, 63 of the 65 longwall faces in the United States were

two-leg shield systems, compared to only 53% in 1985. Larger

size hydraulic cylinders have been developed in recent years to

accommodate the increased demands for higher shield

capacities and have allowed these capacities to be realized with

two-leg designs that were not possible 10 years ago.

There are several consequences of the two-leg design. From

a support strata interaction perspective, the two-leg shield

provides an active horizontal force toward the coal face due to

inclination of the leg cylinders. This active horizontal force

improves overall strata stability by arresting slippage along

fracture planes or by preventing the expansion of highly jointed

or friable immediate roof geologies, which may be further

damaged by the front abutment loading. In terms of the shield

loading, this increase in active horizontal loading also translates

into proportionally higher lemniscate link loading. Historically,

lemniscate link pins and wear in bores have been a primary

cause of premature shield retirement and/or rebuild. Thus, pin

diameters and bore areas need to be increased in higher capacity

shields to prevent this problem from becoming more prevalent.

Another issue related to the two-leg concept is higher contact

pressure on the canopy and base. High toe loading, caused by

the moment created by the line of action of the resultant vertical

forces acting on the canopy and base, can be a problem in highcapacity two-leg shields and should be considered in the

support design. Base toe pressures of $600 psi can be expected

on high-capacity two-leg shields. Base toe lifting devices are

now standard on most two-leg shields to assist in the

advancement of the shields, particularly in soft floor conditions.

There has also been a trend toward solid base designs to reduce

floor-bearing pressures in two-leg shields.

SIZE

In addition to wider shields required by larger diameter leg

cylinders, there is a trend toward wider shields to minimize

hydraulic cylinder maintenance and reduce the total cost by

employing fewer leg cylinders on a longwall face. Again the

issue of torsion (twisting of canopy and base) is important in

designing wider shields. It is not as easy as simply extending

the width of the canopy and base. These components also need

to be strengthened for torsional loading. Additionally, there is

the issue of weight. The 2-m-wide designs may represent an

upper limit with current shield construction materials. If highcapacity designs are to prevail, then lighter weight materials

such as composites are likely to be needed to develop widths

much beyond 2 m. The application of composite materials to

longwall shields is an area that has not yet been explored and

will require a whole new set of engineering requirements to implement into shield design. Although these materials offer significant strength-to-weight ratio advantages, there are problems

regarding abrasion, pin-bearing areas where components rotate

relative to one another, torsion and shear stress control, and

costs.

In addition to increases in width, shields have increased in

length to accommodate one-web-back operations, larger face

conveyors, and deeper shearer webs. Longer canopies and

bases create much larger bending moments that require stiffer

and stronger components than in previous generation supports.

The increase in length is largely responsible for the need for

greater shield capacity as the area of roof loading carried by the

shield increases or the greater convergence is seen by the cantilevering of the immediate roof beam as the resultant shield

force moves further from the coal face.

SETTING FORCES

Setting forces have increased in proportion to the increase in

yield capacity because the size (diameter) of the leg cylinders

has increased to accommodate the higher yield capacities, while

the hydraulic setting pressures have remained constant in the

4,000- to 4,200-psi range. This design practice, coupled with

the increased stiffness of the higher capacity supports, means

that the higher capacity shields are fully loaded as often as their

lower capacity predecessors (figure 4).

HYDRAULIC COMPONENTS AND CONTROL

SYSTEMS

Both Joy and DBT continue to make improvements in the

electrohydraulic control systems, making them more reliable,

more user-friendly, and easier to diagnose when problems

occur. A description of these systems can be found in Barczak

et al. [1998]. Solenoid-operated valving systems are now
227

becoming standard. Spool valves have been shown to be

superior to ball-and-seat designs, which are prone to contamination problems. In addition, these systems allow the solenoid to be activated upon demand, unlike previous systems that

required the hydraulic feed to be interrupted by a control solenoid. This leads to both quicker and smoother control of support functions.

HYDRAULIC EMULSIONS

For many years, longwall shields have utilized a water/

mineral oil emulsion as a hydraulic medium for the leg

cylinders. The standard system has been a 5% oil/water emulsion. There is a trend toward the use of synthetic fluids. Most

western mines have now switched to "low-treatment" systems

with synthetic oils in concentrations of only 1% to 2%. This

has largely been due to environmental issues imposed by the

Utah Department of Natural Resources. Only one eastern mine

is currently using the low-treatment emulsion system. Fazos,

Inc., (Australia) has experimented with an all-water system.

Although the synthetic oils are environmentally preferred,

they cost significantly more. The synthetic concentrate is about

three to five times the cost of mineral oils. Thus, despite the

lower concentration used in the low-treatment systems where

less than 2% oil is utilized, the overall cost is typically about

50% greater than high-treatment systems using 4% to 5%

mineral oil concentrations. The major disadvantages of the

synthetic low-treatment system is that there is little room for

error. A small drop in the oil concentration can lead to lubrication and acidity problems. Therefore, maintenance of the oil/

emulsion is much more critical than in the high-treatment

systems, where the oil content can be reduced from 5% to 4%

with little, if any, detrimental effects. Bacteria growth can also

be accelerated in very low concentrations of oil emulsions,

which can cause more severe corrosion problems than if there

were no oil at all.

SUMMARY OF SHIELD FAILURES EXPERIENCED BY MINE OPERATORS

There have been numerous shield failures throughout the

history of the shield support. A summary of these is provided

below. Although failures have declined, particularly premature

failures that occur early in the shield life, there are still isolated

cases of premature failures. Shields have a finite life, and

fatigue failures will eventually occur on all shields if left in

service long enough.

HISTORICAL OVERVIEW OF SHIELD FAILURES

Base Failures

Base failures seem to be the most prevalent and usually

occur from fatigue after the support has been in operation for

several panels of extraction. A common failure mechanism is

when the leg socket casting breaks away from the base structure. Formation of this failure is difficult to detect while the

support is in service, as the leg socket is housed deep inside the

base structure and this area usually is full of debris. Once the

leg socket breaks loose, the support quickly becomes inoperable. The bottom plates of the base have insufficient

strength to withstand the leg forces, and the leg cylinder will

literally rip the base apart by tearing off the bottom plate.

Failure of the base structure (plates) can also occur without

the leg sockets failing. The probable failure mechanism is

bending of the base. This is more likely to occur in mine sites

that have very strong immediate floor strata. In these hard floor

conditions, steps in the floor may be left by the shearer, as it is

difficult to maintain a constant height of extraction from cut to

cut. The base structure is then simply supported in two

locations and is flexed as loading is applied. Repeated flexure

causes the base to deform (plastically) or promotes fracture

from fatigue, which eventually results in failure of the base

structure. In softer floor conditions, the strata deform to provide a fuller contact to the base. This alleviates much of the

bending and reduces the risk of failure. Standing the support on

the toe of the base can also result in damage of the base structure. This configuration causes maximum stresses in the toe region, and the base will deform or fail usually where the cross

section is a minimum in the section of the base forward of the

leg connection.

Internally, the base structure is constructed with stiffeners

that hold the top and bottom plates apart to form a beam

arrangement, which gives the base its bending strength. Cases

have been reported where these stiffeners have not been

properly welded in place or where the dimensional tolerances

were not within specifications. In these cases, the stiffeners

broke loose and the base structures literally collapsed. This

problem seems to be largely a matter of quality control, but it is

critical to support safety. Since the stiffeners are hidden inside

the base structure, it is virtually impossible (excluding x-ray

inspection) to see these deficiencies prior to the failure.

Canopy Failures

Canopy structures are constructed of stiffened (top and

bottom) plates similar to base structures, and thus they are

susceptible to bending-induced failures as well. Structurally,

canopies are less stiff than bases, which makes them more

susceptible to failure from bending than base structures.

However, while permanent deformation of the canopy is a fairly

common occurrence, destructions of canopies seem to be less
228

Figure 5.—Age distribution of shields operating in 1999.

frequent than observed destructions of bases. This suggests that

canopies are less often subjected to critical bending. Three

reasons why canopies might avoid critical loading are:

(1) immediate roof strata are usually partially fractured, and full

contact with the canopy is more easily obtained, which minimizes bending moment; (2) tip loading on the canopy is

usually smaller than toe loading on the base since the resultant

force is more likely to be located near the toe of the base than

the near the tip of the canopy; and (3) the canopy surface area

is larger than the base area, which allows the canopy to

distribute load more efficiently.

Another common deformation of the canopy is "wrinkling"

of the top plate between the internal stiffeners. This is due

probably to concentrated loading at locations between the

stiffeners, but might also be an indication of failure of the

weldments that hold the stiffeners in place. If the stiffeners are

not secure, the plate may buckle from excessive stress that

results in bending of the plate between the more rigid stiffeners.

Caving Shield And Links

Link members have become considerably more robust in

shield designs over the past 10 years, and failures have been

substantially reduced. Since the caving shield-link assembly

has very little vertical load capacity (stiffness), links are not

highly stressed for most load conditions. Almost all link failures can be attributed to conditions or operating practices that

promote standing the support on the toe of the base or conditions that cause large horizontal displacement of the canopy

relative to the base.

Failure of the structure is most likely to occur in the region

near the (pin) hole located on each end of the member. The failure mechanism is most likely crack formation somewhere on the

circumference of the hole from localized high stress development.

The pinholes elongate from continued wear and contact with the

higher strength link pins. This results in point loading of the pins

and high stress development at the contact areas. These failures

are difficult to detect since this area is obscured from view by the

caving shield clevises or base structures. Although link failures

are rare, they can be catastrophic, as the links provide horizontal

stability to the support structure.

Likewise, caving shield failures are fairly rare, but are more

likely to occur than link failures. While links are designed primarily for axial loading only, shield mechanics indicate the

primary loading mechanism for the caving shield is bending and

torsion. Maximum stresses and failure are most likely to occur

in the clevis areas, where pins connect the link members to the

caving shield. Some general yielding by bending deformation

of the caving shield structure may also occur.

Leg Cylinders

Assuming that the face area is sufficiently stable to prevent

bumps (violent outbursts of energy), it is unlikely that leg

cylinders will experience structural failure since they are designed to control loading by hydraulically yielding at specified

pressures. The most common failure associated with leg cylinders is seal leakage. Internal leakage may occur through other

sources, which will be discussed later in this paper.

Another potential failure mechanism for hydraulic leg

cylinders is malfunction of the yield valves, which allows leg

pressure to increase beyond design levels. Usually, the excessive pressure will cause seal leakage, so that it is unlikely

that sufficient pressure will build up to rupture the cylinder

casing. The more common problem is leakage through the

yield valve, which causes unplanned pressure losses during normal loading. This leakage can be caused by dirt or contamination not allowing the valve to seat properly or worn seats,

defective springs that maintain the valve in the closed position,

or the fitting itself may leak due to bad O-rings or seals. 

FAILURE ASSESSMENT OF SHIELDS CURRENTLY

IN SERVICE

Based on an informal survey at longwall mines, the age

distribution of the shields operating in 1999 is shown in figure 5. The range of operating life extends from 0 to 14 years of

service. An average shield operates about 4,000 cycles per

year; this translates into a face production of approximately

2.5 million tons per year. Thus, an average 10-year-old shield

would have 40,000 loading cycles, whereas a shield employed

on a high-production longwall (4 million tons per year) would

have about 65,000 loading cycles. Approximately one-third of

the shields have been in operation from 4 to 6 years (16,000-

24,000 cycles), and slightly more that than one-third (37%)

have been operating for more than 6 years (24,000 loading

cycles).

Figure 6 shows a near-linear increase in structural failures with

age. Failure is defined as structural damage to any support component. It does not necessarily mean catastrophic failure that

renders the support inoperable. The linear relationship suggests

that more than just fatigue failures are occurring, since
229

Figure 6.—Distribution of structural and hydraulic failures on

shields as a function of age.

the frequency of fatigue failures should increase at a much greater

rate as the shield ages. There were a wide range of structural

failures reported. Most of the major failures involved either the leg

sockets or pin joint problems (i.e., lemniscate link pins and

clevises). These two failures are most likely to put a shield out of

service and/or require structural modification to keep the support in

service. Other structural failures occurred at areas of high stress

concentration, including the side shield on the canopy structure, the

canopy capsule tilt cylinder bracket, the base bridge, the link cutout

areas in the caving shield, various mounting brackets including the

pilot-operated/yield valve manifold, and dishing of canopy skin

covering plate. There were a few isolated cases of leg cylinder

casing problems, but no catastrophic failures.

There were more hydraulic problems reported than structural

problems (figure 6), particularly in shields that have been in

operation less than 6 years. There were nearly four times the

number of hydraulic failures in shields operating 3 years or less

and nearly twice as many for shields in the 4- to 6-year range than

there were structural failures for the same age group. Obviously,

the weak link in shield design is the hydraulic system. Numerous

hydraulic failures or problems were reported. The high frequency

of failures in the first 3 years of operation is distorted by failure

of a component called a hydrafuse. The hydrafuse is a safety

device incorporated by Joy as part of the leg cylinder design on

its recent generation of shield supports in response to a fatality

that occurred recently in Australia. An Australian mine worker

was killed when the hydraulic fluid in the retract annulus of the

leg cylinders was inadvertently pressurized while setting the

support. Normally, the retract port is open through the return line

back to the hydraulic reservoir whenever the leg cylinder is being

pressurized. This is necessary since the pressure would be highly

intensified (by a factor 30 or more) if the port was blocked, as in

the case of the Australian fatality. The resultant pressure

intensification caused the cylinder end cap to rupture, striking and

causing a fatal injury to the mine worker.

The hydrafuse is incorporated into the retract circuit and acts

as a yield valve to prevent the inadvertent buildup of pressure

in the retract annulus of the leg cylinder. The design utilized by

Joy features a brass clip that shears off at a predetermined load

(hydraulic pressure). Once the clip is broken, the "fuse" must be

replaced. With a failure pressure of nearly twice the pump operating pressure, it was expected that the fuse would provide the

desired safety on the rare occasions when there was abnormal

pressure and that replacement of the hydrafuse after the failure

would not be an inconvenience. Surprisingly, hydrafuse failures

began occurring on several shields at several mines. The failures

seemed to occur whenever the support had set idle for some time

and whenever the leg pressure was at or near yield pressure and

almost always on the top stage retract circuit. The National

Institute for Occupational Safety and Health (NIOSH) conducted

a series of tests on a longwall shield under of variety of loading

conditions in the Mine Roof Simulator to determine if unexpected

pressures were occurring in the retract circuit. None were found.

Similar measurements of retract pressure on active shields in

underground mines that reported hydrafuse failures were taken by

Joy; again, no abnormal pressure spikes were found. However,

it is believed that a water-hammer effect is being created by the

valve operation. Joy has since installed a restricter valve in the

circuit to dampen out any potential pressure spike. Early indications from the field installations indicate that this may be

working to prevent the undesirable failure of the hydrafuses.

Unfortunately, the restricter also increases the time required to

lower the top stage. Fortunately, in two-stage leg cylinder designs, the bottom stage is lowered first, and lowering of the top

stage is only required when the bottom stage is fully collapsed.

Thus, the increased time to lower the top stage will not be a

problem during most production mining.

Other cylinder problems caused pressure losses that limited support capacity. One example was due to poor fabrication where the

internal bore of the top stage was off center, resulting in a

weakened casing that was unable to sustain the pressure intensification that occurred in the top stage. Other problems were

reported with defective yield valves that would not reset, resulting

in the inability of the shield to adequately hold load after yielding.

Similar problems were reported with the staging valve, which also

caused loss of pressure in the leg cylinders at a few mines. One

mine reported clearance problems with the leg cylinder and the

base rib plates where the cylinder leaned into the rib plates, causing

internal damage to the cylinder and seal leakage. Failures also occurred with the advance ram cylinders, including failure of shuttle

valve springs and structural failure of a hollow tube relay bar

design.

Most hydraulic problems are related to internal leakage due

to seal wear and/or corrosion of the cylinders. These problems

typically begin when shields have been in service 4 to 6 yrs. As

figure 6 shows, 60% of the shields had some sort of hydraulic

problems during this timeframe. Many problems go undetected

for extended periods of time, resulting in degraded support capacity that can contribute to ground control problems in heavy

loading conditions. Methods to detect the onset of internal

hydraulic leakages are discussed later in this paper. As

previously described, there is a trend toward the use of low-
230

treatment synthetic fluids in western U.S. mines. One western

mine reported severe leg cylinder corrosion due to problems

with a low-treatment synthetic emulsion on shields that have

been in operation less than 3 years.

Problems with the electrohydraulic control systems were also

reported. These included sticking solenoid valves (problem

with soaping in emulsion formulation) and chattering valves

due to fluid dynamics at high flows. Most mines reported that

significant improvements have been made in the latest generation of electrohydraulic controls. In particular, the DBT

(MTA) PM-4 system seems to be a significant improvement

over the previous PM-3 design.

ENGINEERING ASSESSMENT OF SHIELD DESIGN AND FACTORS THAT CAUSE

STRUCTURAL FAILURES

Structural failure can be divided into the following four basic

types: (1) general yielding or excessive plastic deformation;

(2) buckling or general instability, either elastic or plastic;

(3) subcritical crack growth (fatigue, stress-corrosion, or

corrosion fatigue), leading to weakening of the component or

unstable crack growth; and (4) unstable crack extension, either

ductile or brittle, leading to either partial or complete failure of

a member. Structural shield failures are generally of types 1

and 3 and occasionally type 4. A better understanding of the

cause of these failures can be obtained by reviewing the

principles of engineering mechanics for structural design.

BASIC ENGINEERING PRINCIPLES

IN SHIELD DESIGN

Classical structural engineering design is based primarily on

a strength of materials approach. The goal of this approach is

to define the load requirements for the structure and then to

proportion the sizes and shapes of the various components to

prevent tension failure or to prohibit instabilities such as

buckling in a compression failure. Failure is prevented by employing safety factors to ensure that the allowable stress in

components remain within the elastic range of the strength of

the steel. These basic engineering principles are used in current

shield design, except much higher allowable stresses are

permitted in shields compared to the safety factors employed in

the conventional design of structures, such as buildings and

bridges. Failures by general yielding, such as that shown in

figure 7, indicate that shields have been designed without

sufficient margins of safety to prevent yielding. This is in part

due to the low life expectancy of early-generation shield supports. As the demands for greater life expectancy grew, the

manufacturers were forced to employ more conservative design

approaches. However, even with state-of-the-art shields, it is

safe to say that the margins of safety may still remain below

that of conventional structural design where life expectancies

are much longer and failure of any sort is unacceptable.

Since the longwall face and shield supports are advanced

with each shearer cut, the shield is subjected to repetitive

loading and the potential for fatigue-related failures. Fatigue is

the process of cumulative damage that is caused by repetitive

fluctuating loads. Ductile as well as brittle materials are

susceptible to fatigue failures. Since there is no large amount

of plastic deformation prior to the fatigue failure, even in

ductile materials, fatigue failures appear with little or no

warning. The failure mechanism is considered to be quite

complex, but in general a fatigue crack is initiated at some

microscopic or macroscopic stress riser. The crack itself then

acts as a stress riser to promote localized yielding in the vicinity

of the crack tip. In general, the more severe the stress

concentration and the greater the load fluctuation, the shorter

the time to initiate a fatigue crack. After a certain number of

load fluctuations, the accumulated damage causes propagation

of a crack or cracks, leading to failure of the structure.

Conventional structural design for things such as machines

and bridges employ relatively conservative design practices

relative to fatigue loading to ensure the desired life and safety of

the structures. Typically, the fatigue (endurance) limit of conventional steels is about 50% of the tensile (yield) strength. Generally speaking, this means that the nominal load (stress)

developed in the shield components must be kept below 50% of

the yield strength to prevent fatigue failures from occurring,

thereby providing an indefinite life expectancy relative to fatigue

failure. Such a design approach for shield supports would require

much larger component sizes, which would significantly increase

the weight of the shield and, as such, are considered impractical

for shield design. As the stress is increased beyond this endurance limit, the number of cycles to failure is reduced, starting

at about 1 million load cycles. For a life expectancy of 100,000

cycles, which is more representative for longwall shields, the

allowable stress to prevent fatigue failure is much closer to the

yield strength of the steel, suggesting that a factor of safety of 2

is not needed in shield design, at least in relation to fatigue issues.

Of course, these generalities are subject to the specific properties

of the steel and the support construction. The allowable stress

may be considerably less in some circumstances. For example,

the allowable stress to prevent fatigue failure for a plate welded

on the flange of an I-beam (similar to a plate welded on the

bottom of the base pontoon) is only 15 ksi [American Institute of

Steel Construction 1980].

In addition to fatigue, stress corrosion causes failures to

occur under statically applied loads with stress developments

well below the yield strength of the material and independent of

the number of load cycles. This can account for structural
231

Figure 7.—General yielding of toe of base unit in shield design

from early 1980s.

Figure 8.—Corrosion of a lemniscate link clevis.

Figure 9.—Fracture mechanics principles regarding material

toughness, stress, and flaw size.

failures that occur at low-production mines with aging shields

where the number of cycles is low, but the age is high. In other

words, a shield does not necessarily have to have many cycles

to fail. The failure occurs due to crack initiation caused by the

corrosion and subsequent propagation of this crack or material

defects similar to those observed for fatigue loading. The

author believes that stress corrosion is a significant and the

most overlooked factor in current shield design. The mine

environment is often wet and acidic, which accelerates corrosion of shield components, as shown in figure 8. The base

units in particular are subjected to extensive corrosion since the

wet debris continually forms around and often covers much of

the base structure. An area of particular concern is the leg socket, which collects rock and coal debris throughout much of the

shield's operating life. When combined with fatigue loading

(corrosion-fatigue), stress corrosion causes a further reduction

in the useful life of the longwall shield. Corrosion-fatigue damage occurs more rapidly than would be expected from the individual effects or from the algebraic sum of the individual

effects of fatigue, corrosion, or stress-corrosion cracking.

The science of fracture mechanics was developed during

1946-66 to analyze unexplained failures in several large-scale

complex structures where brittle fractures were observed at stress

levels no larger than were expected when the structure was

designed [Barsom and Rolfe 1987]. The underlying premise in

fracture mechanics is that real structures contain numerous

discontinuities of some kind. These discontinuities can be flaws

in weldments, poor fabrication practices where torch cutting

leaves surface scars and abrasions, or cracks in base or weld

materials due to factors like corrosion. These discontinuities act

as stress risers, and unstable, rapid fracturing occurs when the

stress intensity factor at the crack tip reaches a critical value.

Such discontinuities certainly exist in longwall shields and are

undoubtedly a primary source of structural failures.

The goal in fracture mechanics design is to keep the stress

intensity level below the critical value that promotes crack

propagation (figure 9), much like the goal of strength of

materials design is to keep the design stress within the

material's elastic range. To ensure that a structure does not fail

by fracture, the number of cycles to grow a small (often

microscopic) crack to a critical crack length must be greater

than the life of the structure. Thus, the key idea to consider in

a fracture mechanics design is the ability of the steel to absorb

strain energy. If the material has a high absorption capability,

such as mild steel, crack growth will be limited to plastic flow

of the material, and general yielding will typically prevent

brittle fractures from occurring. Conversely, the probability of

fatigue-related failures increases with low-energy absorption

materials. Unfortunately, from a fracture mechanics perspective, state-of-the-art shields are now constructed from highstrength steels (100,000- to 120,000-psi yield strength), which

are much more susceptible to brittle fracture than the early

generation of shield supports, which were constructed from
232

Figure 10.—Reduction in lemniscate pin contact area for worn joint.

Figure 11.—Racking of pin joint.

mild steel (35,000- to 45,000-psi yield strength). The motivation for the high-strength steel, of course, is to minimize the

size of the components and keep the weight of the shield down,

which is an ever increasing challenge for the design engineers

as the capacity of the shields continue to rise. In essence, the

term high-strength steel is a misnomer for two reasons. First,

the fracture toughness is reduced, which means that the

allowable stress, as a percent of the yield strength, must also be

reduced to prevent fracture from occurring. In other words, the

full advantage of increased strength of the steel may not be

realized. Second, unlike lower strength steels where the weld

material is of equivalent or higher strength of the base metal, it

is not uncommon for weaker weld materials to be used in the

high-strength steel constructions.

DESIGN PRACTICES TO IMPROVE STRUCTURAL

MARGINS OF SAFETY AND EXTEND SHIELD LIFE

Since most structural shield failures can be attributed to

some form of fracture, the basic elements of fracture control can

significantly improve shield life. These are: (1) use a lower

design stress, (2) minimize stress concentrations, (3) reduce

flaw size or control crack growth, (4) minimize corrosion, and

(5) use materials of improved toughness. 

Lower Design Stress

Some margin of safety should be employed in the design

stress relative to the yield strength of the steel. Civil engineers

typically use a factor of 1.66, which means that the allowable

stress is about 60% of the yield stress for nonfatigue loading

and further reduced by 50% or more when fatigue loading

applies. While these levels of safety are not practical in shield

design due primarily to cost and weight limitations, it is

important to recognize that a small reduction in (tensile) stress

developments will significantly reduce crack growth since the

two are related by an exponential function. Past practices of

designing to or near yield strength should be avoided in modern

shield design where the life expectancy exceeds 50,000 loading

cycles.

Link pins and clevises are a prime example of historically

poor design practices in shield supports, which continue even

today. Deformation and/or excessive wear in the pin clevises

is undoubtedly the primary cause of premature shield retirement

and/or structural rebuild. There are clear indications that these

areas are subjected to stress beyond the yield strength of the

steel. This poor design is caused partly by manufacturers not

giving sufficient credence to the conditions that cause high

loading in the caving shield-lemniscate assembly, namely, loss

of frictional contact at the roof and floor interface and standing

the support on the toe of the base. With the possible exception

of shields designed for low-seam heights, there is adequate

space available to increase the bearing area of these clevises and

pin diameters to reduce the stress and substantially improve the

life expectancy of these components. The joint design problem

also needs to recognize the importance of pin tolerance. First,

the pin contacts only a portion of the clevis. Typically, arcs of

45E to 60E are used in the design analysis. Obviously, the 45E

arc assumption will lead to more conservative designs.

Conservative assumptions should be made to allow for reduced

areas as wear occurs (figure 10). Excessive pin tolerances can

lead to point loading by allowing the pin to rack within the joint

clevis, as shown in figure 11. Corrosion effects are also often

ignored or not sufficiently accounted for in the design of pins,

despite the fact that corrosion is a leading cause of premature

pin failures or abnormal wear. Corrosion causes pits to occur

in both the pins and clevises (figure 8), which can reduce the

bearing area by 25% to 50% and cause a proportional increase

in stress.
233

Figure 12.—Modification made to strengthen base in order to

prevent leg socket casting failures.

Several failures have occurred where the leg socket casting

weldments break because of fatigue or stress-corrosion and thereby cause failure of the bottom base plate as the loading is transferred fully to the bottom plate instead of being distributed to the

side rib plates. Most of the time, the leg cylinder and casting

punch through the bottom plate and into the mine floor, rendering

the support inoperable. A common modification to alleviate base

leg socket failures of this nature is to add another plate to the

underside of the base pontoon, as shown in figure 12. Typically,

this plate is about 1 in thick and usually covers most of the length

of the base pontoon. Reinforcement is also typically added to the

top area of the side base plate. While this plate stiffens the side

plate, its primary purpose is to restore the location of the

centroidal axis, which was changed by the addition of the bottom

plate, to its location in the original base design. Because of past

failures of this nature, support manufacturers are now beginning

to incorporate a thicker bottom plate in the initial shield design.

Some mines have successfully reduced operating stress

levels by derating the shield support before underground installation. The derating is accomplished by installing yield

valves with a lower operating pressure than specified in the

design. Since the leg capacity controls the maximum load developments within the support, lower leg loads translate into

reduced component loading. For example, if a 1,000-ton shield

support is derated by 10% to a 900-ton capacity, the margin of

safety relative to the tolerable crack size that will prevent

fatigue fracture may increase by 20% to 30%.

Minimizing Stress Concentrations

There are numerous sources of stress concentrations in longwall shield supports. The most common is a change in geometry. These stress concentrations should be identified and

their magnitudes quantified during the design and performance

testing phase. One way to do this is to use photoelastic plastics.

The photoelastic plastic can be applied to almost any area of the

shield structure. Colored fringes will appear on the plastic

when observed through a polarized lens that correlate to the

stress profiles. Stress intensity factors of 2 to 3 are not uncommon for sharp changes in geometry, such as holes or sharp

bends in structural members.

One example of a sharp change in geometry is a lemniscate

link design with offset pinholes, as shown in figure 13. This

link design is typically employed on a shield with a low profile

designed for operation in low-coal seams. The bend is necessary to provide clearance with the caving shield in the

collapsed or low operating height. Lemniscate links are primarily axially loaded members, but the offset pinhole geometry

induces additional stresses due to bending and thereby significantly reduces the margin of safety for this component.

Figure 13 depicts failure of a bottom lemniscate link on a 620-

ton Westfalia shield that occurred during performance testing

at NIOSH's Safety Structures Testing Laboratory. Although the

shield had 45,000 load cycles from underground service before

testing, a new link was installed on the shield that was

performance tested in the laboratory. This was a new link

design that was fabricated for the mine by an outside vendor

(not the support manufacturer). Failure occurred after only

14,000 loading cycles. This failure illustrates two problems.

First, the design (allowable stress) was too high. Test results

revealed that the nominal stress in the link exceeded 90% of the

material yield strength. In addition, it appeared that the link

side plates had been torch cut, adding an additional stress riser

to the already sharp change in geometry at the bend in the link

in the area where the failure occurred. These two factors resulted in an unacceptable time to failure. Figure 14 illustrates

another problem where the fabrication process left large flaws

that led to premature fatigue failure of the base rib. This

problem may have been alleviated if the surface had been

smoothed to remove most of the surface flaws.

Any hole in a structural plate is another area where stress is

concentrated. The structural components of a longwall shield

(canopy, caving shield, lemniscate links, and base) are connected

by a pin and clevis arrangement. These areas are also sources of

stress concentration and fatigue failures on aging longwall shields

(figure 15). Another example of a stress concentration caused by

a sharp change in geometry is shown in figure 16. Holes are

sometimes cut into the canopy or caving shield structure to

accommodate the placement of hydraulic hoses. In the example

shown in figure 16, failure occurred at the stress riser caused by

the sharp corner in a cutout on the caving shield made to

accommodate hosing for the side shield. The best solution to this

problem is to avoid the hole altogether, and if the hole is

necessary, to ensure that the corner radius is as large as possible.

Another example of failure due to stress risers created by sharp

geometries is shown in figure 17, where a crack developed in the

caving shield near the canopy hinge. A clean-out hole is often

placed in the side of the base structure to facilitate removal of

debris from the leg socket area. This too can be a source for concentration of stress. Since this is a critically loaded area of the

base, care should be taken in the design to minimize the stress

concentration by a incorporating a favorable geometry and orientation with respect to the stress field in this member.
234

Figure 13.—Failure of bottom lemniscate link.

Figure 14.—Failure of base rib side plate where fabrication

process left large flaws.

Figure 15.—Failure in base lemniscate link clevis.

Figure 16.—Failure due to stress concentrations in cutout

sections of caving shield.

Figure 17.—Failures of caving shield due to stress concentration in sharp corners where lemniscate links connect.

Figure 18 shows failure of a canopy leg socket casting due

to a stress riser created by a sharp change in geometry. The

failure in this case occurred in the casting itself. This failure

occurred on several shield supports in the late 1980s, all of

which used this same basic socket design. Again, this failure

probably could have been prevented by a smoother geometry.

Leg sockets are often a source of fatigue failures in aging

longwall shields. The leg socket is a critical area since the full

load developed within the hydraulic leg cylinder must be

transferred into the canopy and base structures to be distributed

to the mine roof and floor. An examination of the structural

mechanics associated with these base socket failures has led to

some other design changes that are intended to reduce the stress

concentration in the leg socket casting/base structure connection. The rectangular geometry of the socket casting and its

placement between the side rib plates of the base structure
235

Figure 18.—Failure due to stress riser created by sharp corner

in canopy leg socket casting.

Figure 19.—"Zipper-shaped" leg cylinder casting to avoid

stress concentration caused by orientation of casting to principal

stress field.

position it at right angles to the principal tensile stress caused by

the bending of the base structure. This orientation creates a stress

intensity factor, which helps to promote fatigue-induced fracturing of the weldment. Some shields are now being designed

with an elliptical or zipper-shaped casting (figure 19) so that the

front and/or rear edge is not perpendicular to the principal tensile

stress, thereby resulting in a reduction of the stress intensity

factor.

Reducing Flaw Size or Controlling Crack Growth

Weldments are an essential part of the shield fabrication process.

However, since weldments are a primary source of structural flaws,

the quality of the welds is critical. As the principles of fracture

mechanics illustrate, the initial flaw size created by the welding

process is critical to the crack propagation and the margin of safety

achieved in this structure. Once a crack develops, it is desirable to

keep the crack contained within the weld and not have it progress

to the base material adjacent to the weld. This action may depend

on the nature of the heat-affected zone in the immediate vicinity of

the weld. In general, the heat-affected zone results in anisotropic

material properties and residual stresses that tend to reduce the

toughness of the steel in this area and increase the likelihood of

fracture into the base metal. For the high-strength steels used in

modern shield supports, proper heating and cooling of the steel

during the weldment process are crucial to preventing crack

initiation and growth in the heat-affected zone. This is why it is

very difficult to conduct repairs to damaged shields underground,

since it is virtually impossible to be able to preheat and properly

cool the steel when welding at the longwall face. Another approach

that can be used to keep crack growth contained to the weldments

is to create breaks in the weldments. This practice is sometimes

used in leg socket castings. The break in the weld at the corners of

the casting acts like as a crack arrester to stop the growth of the

crack. This same technique is used in the airplane industry by

drilling holes in the metal at the end of an observed crack before it

reaches a critical crack length.

Perhaps the best approach to avoid problems associated with

weldments is to eliminate them when possible. A good example of this pertains to the leg socket design. A typical construction for the leg cylinder base socket is shown in figure 20.

A casting 3-4 in thick and 18-24 in long with a spherical seat to

accommodate the bottom of the hydraulic cylinder is placed on

top of the bottom cover plate on the base pontoon and is welded

in place along the four sides of the top of the casting to the side

rib plates and cross plates that connect the two side base plates

together. This design is highly dependent on the welds to

transfer load into the side rib plates and maintain the structural

integrity of the socket connection. An alternative design that is

now being used in some canopy leg sockets to alleviate the

weld fatigue problem is to cut rectangular holes into the side

base rib plates and extend the width of the casting so that it

bridges across the cylinder opening, but is supported by the side

rib plates (figure 20). In this configuration, leg cylinder loading

is transferred directly to the side rib plates of the base structure

entirely through base metal contact and is not dependent on the

weldments to achieve this load transfer.

Corrosion Control

Some steps can be taken to circumvent the problems caused

by corrosion. The most convenient approach is to try to protect

the shield from the environment. For shield supports, this applies mostly to painting with an industrial paint that is resistant

to the wet mine environment. However, for some components

such as the pins and clevises where there is considerable wear

due to the kinematics of the shield during load application,

painting or even plating with more resistant material is not an

option. Sherardizing pins has proven beneficial, but the effectiveness is limited due to the wear in the joint. Another option

worth considering is some form of lubrication for these joints.

A simple grease fitting would be an improvement, although

somewhat impractical to maintain on an active longwall face.

Thus, a sealed joint of some form would be preferable.
236

Figure 20.–Conventional and alternative leg cylinder casting

design.

Tests have shown that compressive stresses on the surface of

materials that are exposed to the environment will not

necessarily minimize corrosion-fatigue crack initiation, but they

can reduce the possibility of crack growth and thereby prevent

failure from occurring. This could be done by induction

hardening the joint pins. This, however, might make the pins

more brittle.

Material Toughness

The strength of materials approach essentially ignores the

toughness of the material, which as described is the most

significant material property for fracture control. There is a

tradeoff in the use of high-strength steels if they provide

superior strength but reduced toughness with greater chance for

failure. Since the author is not familiar with the German and

U.K. steels used in the shield construction, no specific

recommendations are made here, but the issue should be

investigated with the shield manufacturer when new shields are

purchased.

 KEY POINTS IN EVALUATING HYDRAULIC FAILURES

As described earlier, hydraulic failures are common on

aging longwall shields. Although hydraulic failures are

generally not as catastrophic as structural failures, they cause

a reduction in supporting capability that can lead to serious

ground control problems.

FUNDAMENTALS OF SHIELD HYDRAULICS

Hydraulic Cylinder Operation

Since the capacity of a shield is controlled by the hydraulic

leg cylinders, a basic understanding of their operation is

essential to understanding shield design and causes of hydraulic

failures [Barczak and Gearhart 1998]. The basic operation of

a hydraulic cylinder can be described as follows (see figure 21).

A hydraulic power supply pumps fluid into the cylinder cavity.

The fluid acts against a piston, causing the piston and attached

steel rod to displace outward. This displacement will continue

with very little hydraulic pressure until the support is set

against the mine roof and floor. The hydraulic pressure then

increases rapidly until the full pump pressure is reached. After

the pump supply is turned off or isolated from the support, this

pressurized hydraulic fluid is trapped inside the cylinder by a

pilot-operated check valve. The pressure of the hydraulic fluid

inside the cylinder is then intensified in proportion to any

increase in roof loading. A yield valve limits the maximum

pressure inside the cylinder to prevent excessive loading that

would damage the cylinder. In longwall shields the cylinders

are double acting, which means that they are hydraulically

powered to both extend and retract (figure 21). Powered

lowering of the support is achieved by pumping fluid inside the

retract annulus (figure 21) while at the same time applying pilot

pressure to open the check valve to allow fluid from inside the

main cylinder cavity that supports the weight of the mine roof

to escape back to the hydraulic power supply tank.

The previous example describes the basic operation of a

single-stage cylinder, where extension and retraction is provided by one stage. Longwall shields utilize two-stage hydraulic cylinders, as shown in figure 22. Incorporating more

stages into a leg cylinder design generally allows for a lower

collapsed height, combined with an equal or greater maximum

working height to provide operation in a wider range of mining

heights.

The basic operating principles of these multistage designs

are the same as the single-stage hydraulic cylinder previously

described, but there are some noteworthy differences relative to

how the individual stages perform:

1. The bottom stage extends and retracts first, followed by

the top stage. The bottom stage will extend until it is fully

stroked before the top stage will begin to extend.

2. A check valve is installed in the bottom-stage piston

(figure 22). This check valve functions to allow hydraulic fluid

to flow from the bottom stage to the top stage to cause the top

stage to extend during the setting operation whenever the

bottom stage is fully extended. It also isolates the bottom stage

from the top stage to allow the hydraulic fluid to be intensified

in the top stage once the support is actively set against the mine

roof and floor. This is necessary to allow the top stage to carry

the same load as the bottom stage.
237

Figure 21.—Functional diagram of hydraulic cylinder operation.

Figure 17.—Failures of caving shield due to stress

concentration in sharp corners where lemniscate links

connect.

Understanding how the staging functions to provide the

necessary support capacity is critical to a proper understanding

of multistage leg cylinder operation. For proper operation of

the support to occur, force equilibrium must be satisfied for

each stage, which means that each stage must carry the same

load. This requires different hydraulic pressures in each stage,

since the area of each stage is different. Mathematically, this

requirement is expressed by equation 1. If the top stage area is

one-half that of the bottom stage area, the pressure in the top

stage will be twice that of the hydraulic fluid in the bottom

stage.

A1 ( P1 ' A2 ( P2, (1)

where A1 ' area of largest diameter (bottom) stage, in2

;

P1 ' pressure in largest diameter (bottom) stage, psi;

A2 ' area of the top stage, in2

; and

P2 ' pressure in the top stage, psi;

The extension of specific stages at a particular point in time

(cycles of operation) depends on the history of the operating

heights of the support. When the support is initially raised from

a fully collapsed position, the bottom stage will extend first. If

the mining height is greater than the stroke of the first stage, the

top stage will extend until roof contact is made. On subsequent

cycles, the top stage will remain at the initial extension until

(1) the mining height is increased beyond that of the initial

cycle, or (2) the support is lowered such that the bottom stage

is fully collapsed. Assuming that the support is not lowered to

the point where the bottom stage is fully collapsed, the top stage

will extend beyond the initial extension only when the mining

height is increased. In essence, the bottom stage will be fully

stroked (1) when the support is initially raised from a collapsed

position and (2) on the mining cycle that establishes a new

maximum operating height, whenever the support height is

higher than it has been on all previous mining cycles. On all

other cycles, the bottom-stage extension will reflect changes in

mining height relative to the initial mining height and the

maximum operating height at which the support was utilized.

An example is used to illustrate these concepts. As shown

in figure 23, the initial setting of the support causes the bottom

stage to be fully extended 24 in (stroke) and a 12-in (partial)

extension of the top stage. If the roof-to-floor mining height is

reduced on the second operating cycle by 3 in, the bottom-stage

extension now will be 21 in, while the top-stage extension will

remain at 12 in. If the mining height is increased by 6 in from

that of the second cycle, the bottom stage will again be fully

extended (24-in stroke) and the second-stage extension will

increase to 15 in on the third cycle. This exercise can continue

as shown in figure 23. As shown in this exercise, the top-stage

extension on any cycle will equal the initial extension plus the

incremental increase in mining height beyond the initial mining

height. The top-stage extension will never be less than the

initial extension unless the support is lowered to the point

where the bottom stage is fully retracted. The bottom-stage

extension will fluctuate with changes in mining height on each

cycle and will be fully stroked whenever a new maximum

operating height is attained.
238

Figure 23.—Example of stage extension for changes in mining height.

Active (Setting) Loads

The shield is actively set against the mine roof and floor by

pressurizing the hydraulic cylinders. The force exerted against

the mine roof and floor is dependent upon the size of the

cylinder and the hydraulic (pump) pressure. In multistage

cylinders, the setting force can be determined by measuring the

pressure in any stage and multiplying it by the (piston) area of

that stage. It is important to remember that the pressure in the

top stage is intensified during the setting operation, and only the

bottom stage will be at pump pressure once the support is set

(assuming that none of the stages are fully stroked). In longwall shields, the hydraulic pressure is measured only in the

bottom stage; thus, the pump pressure times the bottom stage

area should be used to calculate the setting force. The total

setting force for the support is simply the sum of individual

hydraulic cylinder forces.

If the bottom stage is fully stroked, the setting force will be

diminished. When the bottom stage that is fully extended no

longer transfers all of its force to the mine roof and floor, part

of it is consumed by tensioning the cylinder casing as the piston

is forced against the mechanical stops that limit its travel. The

pressure in the top stage will not be intensified, but instead will

be at pump pressure. The setting force will always equal the

pump pressure times the area of the largest diameter stage that

is not fully extended. Thus, the setting force will equal the

pump pressure times the top stage area whenever the bottom

stage is fully extended.

The reductions in setting force due to full extension of the

bottom stages are typically between 40% and 50% for most

shield designs. Thus, a 1,000-ton shield with a setting load of

670 tons at 4,200-psi pump pressure would have a setting force

of only 268 to 335 tons when the bottom stage is fully

extended. Recalling the operation of the hydraulic cylinder

previously described, the setting force will be diminished

whenever the bottom stage is fully extended, which occurs on

the operational cycle where a new maximum operating height

is attained. The setting force will be restored to its full capability on all other cycles or whenever the support is operated at

a height less than the highest operating height on any previous

operating cycle.

Yielding Behavior

Yield valves are normally connected to the bottom stage of

multistage hydraulic cylinders to provide overload protection.

These valves open whenever the pressure exceeds the design

threshold and allow the pressurized fluid to escape from the

bottom stage of the cylinder. This loss of fluid causes the

pressure to drop in the bottom stage until the valve reseats. The

reseating pressure is typically about 90% of the yield pressure;

thus, the support capacity will drop by approximately 10%

when the yield valve opens. As the pressure drops in the bottom stage, a force imbalance occurs between the top and bottom

stage, which causes the bottom stage to lower until force

equilibrium is attained. The top-stage extension does not

change during yielding until the bottom stage is fully collapsed.

How much the bottom stage lowers because of yielding

depends on three factors: (1) the area of the bottom stage, (2) the

extension of the bottom stage, and (3) the yield pressure and

reseating pressure of the yield valve. The reduction in support

height due to yielding will be more for large extensions of the

bottom stage than for small extensions of the bottom stage,

decreasing in direct proportion to reductions in the bottom-stage

extension at the time of yielding. Thus, on each successive yielding during any one operating cycle, the lowering of the bottom

stage will be progressively less. The magnitude of bottom-stage

lowering due to yielding can be calculated using equations 2 and

3. Using a 700-ton longwall shield with a bottom-stage diameter

of 11.8 in and a yield pressure of 6,389 psi as an example, the

bottom stage of each hydraulic cylinder will drop approximately

0.028 in during a single yield event if the bottom stage was

extended 14 in when yielding began. A 70-ton reduction in

support loading would be caused by the yielding.
239

∆D ' ∆V

A ,

Figure 24.—Imbalance in leg cylinder staging due to internal

leakage.

∆V ' ∆P(V

β , (2)

where ∆V ' change in volume of fluid in bottom stage of

 cylinder, in3

;

∆P ' change in hydraulic pressure in bottom stage of

 cylinder, psi;

∆P ' yield pressure minus reseating pressure, psi;

V ' volume of fluid in bottom stage at time of

 yielding, in3

;

V ' area of bottom stage times extension of bottom

 stage, in3

; and

β ' bulk modulus of oil, psi (i.e., 320,000 psi).

where ∆D ' reduction in bottom-stage extension due to

 yielding, in;

∆V ' reduction in volume of fluid in bottom stage

 due to yielding, in3

; and

A ' area of bottom stage, in2

.

DETECTION OF HYDRAULIC FAILURES

It is important to realize that all shields will experience

leakages due to seal wear and/or component failures several

times during their life expectancy in the mine. Also, many of

these failures will go undetected for extended periods of time,

resulting in significantly degraded shield capacity and ground

control capability. Furthermore, if shield life is to be maximized, then recognition and correction of hydraulic failures are

important. If both leg cylinders are not functioning properly,

eccentric loading in the canopy and caving shield-lemniscate

assembly, for example, will be created, causing increased

probability of structural failure in these components. Additional loading may be transferred onto adjacent shields, thereby

reducing their life expectancy. Degraded support capacity may

induce more severe weighting of the shields, creating a

snowball effect that further degrades the support capability.

Observations of the relative positions of the cylinder staging

can be used to identify cylinder problems and the cause of

hydraulic leakages. One indication of internal leakages is when

the bottom stage is consistently fully extended. The bottom

stage should be fully extended only on operating cycles that

establish a new maximum operating height. Thus, on the majority of operating cycles, the bottom stage should not be at full

extension. Another indication of hydraulic leakage is when

there is a large difference in stage extensions of leg cylinders on

the same support (figure 24). Observation of the bottom-stage

position can help to identify the component failure (table 1).

Table 1.—Stage movements associated with

internal component failures

Component failure Bottom-stage

movement

Staging check valve . . . . . . . . Up.

First-stage seals . . . . . . . . . . . Down.

Second-stage seals . . . . . . . . . Up.

Pilot-operated check valve . . . Down.

Yield valve . . . . . . . . . . . . . . . . Down.

Problems with the staging check valve can be isolated by

fully collapsing the shield and monitoring the leg pressures on

the operating cycle after the support is reset against the mine

roof and floor. The requirement is to have the bottom stage

fully extended when the support is reset. In this configuration

with a staging check valve that is functioning properly, the

pressure in the bottom stage will not change significantly until

the force in the top stage due to additional roof loading

overcomes the setting force developed in the bottom stage

(figure 25). An immediate increase in pressure in the bottom

stage indicates that the check valve is leaking sufficiently to not

allow the pressure in the top stage to be intensified.

Two factors that reduce hydraulic life expectancy are

contamination and corrosion. It is very important to change

filters as needed to maintain a clean emulsion. Debris in the leg

cylinder is a leading cause of stage valves failures, which

reduce shield capacity by 40% to 60%. Debris also reduces

seal life and can cause check valves and yield valves to malfunction. Corrosion is also a primary factor in leg cylinder life.

A chemical analysis of the emulsion fluid should be done

periodically. Bacteria growth or poor water quality can significantly reduce the life of the leg cylinders.

From the design and repair perspective, there have been

improvements in plating technology, which has reduced the

corrosion problem in recent years. Early shield designs used

chrome plating for the leg cylinder bores. Chrome is a very

porous material, and even if the thickness is increased, the

structure, which is analogous to placing layers of chicken wire

on top of each other, remains quite porous. Bronze is a much

better material, and most shield cylinders are now plated with

(3)
240

URL ' A2( P2 & A1( P1

2000 ,

Figure 25.–Identification of defective leg cylinder staging valve.

bronze. The harder the bronze is, the better the life that can be

expected. An alternative to bronze is to use chrome over top of

nickel plating. The electrolysis of the nickel creates a very hard

material that is highly corrosion-resistant, but the process is also

considerably more expensive. Another problem with using

nickel as the base metal in the plating process is that the two

metals tend to react with one another, reducing the bond between them, which causes the chrome to peel off in time. 

An impending problem with current plating technology are

the environmental hazards associated with the plating process.

Both chrome and bronze plating present environmental hazards

that are likely to drive up the cost of these conventional plating

methods in the near future. Since rebuilding of shield leg cylinders is a routine part of extended shield operation on every

longwall, this can have a significant impact on the mining industry, and new plating technologies need to be explored.

Swanson Plating, Morgantown, WV, is now offering laser-controlled plating as an alternative. The laser technology, in addition to being more environmentally friendly, reduces the heat

affected zone by nearly two orders of magnitude and allows for

much more controlled plating using materials that cannot be

applied by conventional plating practices. More innovative solutions to consider would be the use of ceramic or composite

materials that are much more corrosion-resistant than the heavy

metals currently used in cylinder construction.

ERRORS IN ASSESSING SUPPORT LOADING

Support loading is determined from measurement of the

pressure in the bottom stage of the hydraulic cylinder. Since

the pressure increases in direct proportion to the increase in

support loading, this provides an accurate assessment of the

roof loading through the full loading cycle. However, when the

bottom stage is fully stroked, a portion of the increase in roof

loading after the support is set against the mine roof and floor

will not be detected by changes in hydraulic pressure in the

bottom stage. The reason for this period of undetected roof

loading is that when the bottom stage is fully extended, the

bottom-stage piston is being held against the mechanical stops

with a force exerted by the pump pressure at the time the

support was set. The pressure in the bottom stage will only

increase when this piston is moved off of the stops and begins

to compress the hydraulic fluid in the bottom stage. In order for

this to happen, the pressure in the upper stage, which is at pump

pressure when the support is set, must increase to cause a force

in the upper stage that exceeds the setting force in the bottom

stage. The additional roof loading that is required to produce

this additional force in the upper stage is the roof loading that

is undetected by the pressure gauges measuring hydraulic

pressure in the bottom stage.

The mechanics of the pressure development in the various

support stages can be described by examination of equilibrium

requirements for each individual stage, realizing that each stage

must carry the same load. The undetected roof loading with the

bottom stage fully extended can be calculated using equation 4.

An examination of these principles indicates that the period and

magnitude of undetected roof loading will increase as the

setting pressure increases for a particular support.

where URL ' undetected roof load, tons;

A1 ' area of first (bottom) stage, in2

;

A2 ' area of second stage, in2

; and

P1 ' P2 ' setting (pump) pressure, psi.

Figure 26 depicts undetected roof loading for a 700-ton

(two-stage) longwall shield set at full pump pressure. It is seen

(4)
241

Figure 26.–Undetected roof loaded by bottom-stage pressure measurements when bottom

stage is fully extended during setting of the shield.

from this figure that the unrecorded roof loading is quite large.

For example, approximately 200 tons of undetected roof

loading, which represents 84% of the remaining shield

capacity, will occur on the 700-ton longwall after it is set

against the mine roof and floor.

CONCLUSIONS AND RECOMMENDATIONS

The evolution of improved shield design continues with the

current generation of shield supports. The previous generation

of shield supports (1985-95 era) experienced several serious

structural failures. These included canopy sections falling off,

bases cracking in half, leg sockets pushing through the base and

canopy into the mine roof and floor, lemniscate links breaking,

and pins bending and breaking. Generally, these problems have

been addressed by the support manufacturers. However, as the

capacities continue to grow, each new generation of shield supports must also experience the growing pains of design deficiencies. This is occurring now as isolated failures of modern

shield design are again cropping up.

While proper shield design is not magic, there is an

experience factor that cannot be overlooked. Unlike designing

a building or a bridge where the load conditions are well known

and fail-safe design philosophies can be readily employed, the

mining environment is not well defined or easily understood.

Mining engineers typically design to allow failure, while conventional structural engineers design to prevent failure.

Furthermore, as the demand for extended shield life continues

to grow, continued improvements in quality control and innovations in shield design must be developed if this requirement is to be realized.

Although the design of longwall shields is a complex issue,

it would be beneficial to prospective shield buyers to learn the

basic engineering principles associated with shield design

issues that are addressed in this paper. Although it is not necessary to know all of the answers, knowing what questions to

ask is important. Hiring a structural engineering consultant is

another option, but again one must be aware of the uniqueness

of the mining environment and shield design issues.

Corrosion is by far the most overlooked factor in both shield

design and performance testing. Corrosion control practices are

limited, but they need to be followed. Simple things such as

using corrosion-resistant paint and repainting of shields can be

effective. Design changes need to be made to the pin joints and

clevises, which are subjected to high wear rates, to avoid

corrosion in these areas. Advanced plating technologies and

more corrosive-resistant materials need to be employed to

improve hydraulic cylinder life, which is by far the weak link in

a shield's lifeline.

Support failures are site-specific, and the magnitude of the

problem must be considered in making judgments of support

safety. The information provided in this paper will help identify problems and suggest solutions. However, the final

judgment must be made with considerations of the severity of

the problem, number of supports affected, past performance

history, face conditions at the time of failure, and the overall

situation at the mine site. Finally, the best policy is always to

correct problems as soon as they occur. Although this is often

impractical and at times impossible, these generic recommendations regarding safety precautions of problem support

systems are made with the realization that they are not universal

for all circumstances:

1. When failures occur, procedures to reduce shield load

should be implemented.

A good practice to reduce face weighting is to maximize the

rate of advance. Idle faces generally have a tendency to create

higher shield loads. Also, setting forces should be optimized.

Conventional wisdom has been to increase the set pressures or

to ensure that full setting (pump) pressure is maintained. However, in displacement-controlled loading where the load development is proportional to the shield stiffness and the support

resistance is not fully controlling the ground movements,

reducing the setting loads may also be an option to reduce the

total loading on the support. Another possibility is to reduce
242

the depth of the shearer cut. Much of the shield loading is generated from the shearer pass. A reduced cut may reduce shield

loading, although there are no data to substantiate this

hypothesis.

2. Ensure proper operation of the support once it is put into

service.

Another issue related to lowering the design stress is to

ensure proper operation of the support once it is put into service

in the mine. Stress developments are greatly enhanced whenever the support is stood on the toe of the base. When a shield

is set with the canopy tip up, the leg forces will typically rotate

the canopy into full roof contact and in the process lift the rear

of the base off of the ground. This base-on-toe configuration

increases the magnitude of stress in the lemniscate links by as

much as 300%. Modern two-leg shields employ a control system for the canopy capsule cylinder that prevents setting the

shield with the canopy tip up. Thus, if this system is deactivated by the mine personnel, the life expectancy of the pin

joints and the entire shield can be significantly shortened.

3. Recognize the cumulative effect of degraded performance

and noncatastrophic failures.

As previously discussed, it is quite common for aging

shields to experience hydraulic leaks that lead to degraded

support capacity, but typically do not interfere significantly

with the stability of the support or its capability to remain

functional. Whenever the degraded support capacity is accompanied by a lack of ground control, obviously the problems

must be corrected immediately. However, with modern highcapacity supports this is generally not the case, and the question

is how soon the degraded leg cylinder performance should be

corrected. In many cases, leakage will occur in only one leg or

be considerably worse in one leg compared to the other. From

a load distribution viewpoint, loads will transfer down the side

of the structure with the active leg. This imbalance in leg

forces will cause some increase in component stresses, particularly in the caving shield and lemniscate assembly. Although this generally will not pose an immediate threat, the

cumulative effect of the increased loading, if left unattended,

will more than likely decrease the shield life by accelerating

fatigue-related failures. Furthermore, if structural failures do

occur, the asymmetric loading caused by imbalanced leg forces

will likely make these failures more severe.

This same logic applies to structural failures. There is considerable redundancy built into a shield support in the sense that

there are multiple load paths to transfer roof loading through the

support structure and into the mine floor. Most failures when

they first occur will not immediately affect the performance of the

support. However, once a failure initiates, a domino effect will

likely occur wherein additional loading is transferred elsewhere,

and the probability for the failure to grow or spread to other

components as time progresses is also likely.

4. While fatigue failures are difficult to judge, they can lead

to catastrophic failure with little or no warning. 

Modern shields of high-strength steel are more susceptible

to unstable failure than previous generations of shields constructed from mild steel. Crack formation in any part of the

support structure should be viewed as a sign of potentially

imminent danger. The crack indicates that the steel has failed.

Whether this crack will propagate to cause destruction of a

support component depends on many things, most notably, the

ability of the member to effectively redistribute loading primarily within that component or, to a lesser degree, to other

components. This makes judgments of support safety in these

situations difficult. In any event, cracks in a support structure

should be closely monitored, and attention should be given to

any increase in the growth and particularly the rate of growth of

the crack. Cracks that arrest themselves after formation can

often be ignored, particularly when they are in sections of the

support that do not carry the bulk of the loading. In this case,

load was adequately transferred elsewhere to alleviate the

localized stress condition that initiated the crack. Conversely,

a crack that continues to grow after formation is likely to

develop into a critical situation where the safe performance of

the support is threatened. Obviously, as soon as the structural

integrity of the support (component) is threatened, the support

should be taken out of service and modifications made. Once

a critical crack length is reached, the growth rate can accelerate

in relatively few additional loading cycles. Particular attention

should be paid to weldments, since localized stress developments are likely to be higher in these areas and the potential for

rapid crack growth is enhanced.

5. Recognize failures that can lead to instability and correct

them immediately.

It is unlikely that the support will show any signs of diminished load-carrying capability while structural problems are

developing, but the stability of the support may be threatened.

Failures to the caving shield or lemniscate link pins or to the

links themselves are most serious since they control the horizontal stability of the shield. Failure of any set of link pins or

links will cause the support to collapse under its own weight

from instability. There is redundancy built into the shield in

that there are pairs (left and right side) of both the upper and

lower links. Failure of one side may not result in immediate

instability, but will significantly increase loading in the adjacent

link and therefore should be viewed as very serious, with

immediate action taken to remove the failed pin and install a

new one.

Another serious failure that can lead to instability is failure

of the leg sockets. The leg sockets must transfer loading from

the canopy to the base. If the sockets break lose, the cylinder

will eventually punch through the bottom plate on the base and

render the support inoperable and perhaps unstable.
243

6. Understand the hazards of automated control systems.

During the early development of the electrohydraulic control

systems with automated and shearer-initiated advance, several

injuries occurred because of unplanned or unexpected shield

movements. Although failures in the electronics have largely

been eliminated in recent years, occasional injuries still occur

where a miner is unaware of the pending shield move, which

was initiated at a remote position by another miner. Close

attention should be paid to the audible warning of the shield

advance. A good practice is to walk on the bases of the support

whenever possible to avoid being pinched by an unexpected

shield move.

REFERENCES

American Institute of Steel Construction [1980]. Manual of steel

construction. 8th ed. Chicago, IL: American Institute of Steel Construction,

pp. 5-89.

Barczak TM, Gearhart DF [1998]. Performance and safety considerations

of hydraulic support systems. In: Peng SS, ed. Proceedings of the 17th International Conference on Ground Control in Mining. Morgantown, WV:

University of West Virginia, pp. 176-186.

Barczak TM, Chekan GJ, Babich DR [1998]. Shield design, construction,

and operation. In: Proceedings of Longwall USA International Exhibition &

Conference, pp. 5-47.

Barsom JM, Rolfe ST [1987]. Fracture and fatigue control in structures.

Englewood Cliffs, NJ: Prentice-Hall Inc.

Fiscor S [1999]. U.S. longwall census '99. Coal Age Feb:30-35.
245

FACTORS TO CONSIDER WHEN PURCHASING A NEW SET 

OF LONGWALL SHIELDS

By Thomas M. Barczak1

ABSTRACT

Purchasing a new set of longwall shields requires a substantial investment. A poor shield design can lead

to economic hardships, safety concerns for the mine workers, and closure of the mine. This paper addresses

several key points that should be considered in the procurement process: (1) understanding your goals and

the logic in selecting a higher capacity shield, (2) the importance of completing performance testing before

production shield fabrication begins, (3) making sure performance testing is properly done, (4) measuring load

(stress) development during performance testing, (5) testing a shield to failure, and (6) the value of buying

extra shields. In addition, several challenges are proposed for consideration in future shield designs. These

include (1) 100,000 life-cycle expectancy, (2) improved hydraulic diagnostic capability, (3) smart load control

by optimizing setting pressures, (4) lubricated link joint design concepts, (5) composite material applications

to reduce shield weight, (6) incorporating periodic weighting predictive algorithms into the data-processing

software, (7) advanced component load measurement on two to three specially instrumented shields to detect

loading in the structural components, and (8) constant set leg cylinder design.

1

Research physicist, Pittsburgh Research Laboratory, National Institute for Occupational Safety and Health, Pittsburgh, PA.
246

INTRODUCTION

A new set of longwall shields can cost over $20 million.

This represents a major investment for any mine operator.

Currently, these new shields are expected to last for more than

60,000 loading cycles, representing 18-20 panels of mining

[Barczak 1999]. A poor design or even a marginal design that

reduces the operating life of the shield can be the difference

between a mine staying in business or closing. In addition to

the economics pertaining to the initial shield investment, time

and money spent in making repairs to damaged shields can be

very costly.

There are also safety issues when failures occur. Poor shield

performance can contribute to instability in the overlying rock

mass leading to roof falls in the immediate face area, cavity

formation above the shields, or even "iron-bound" shields when

excessive yielding occurs. The well-planned removal of shields

in properly designed recovery rooms is difficult enough. There

are hazards associated with increased exposure of the mine

workers to unsupported ground and with moving very heavy

equipment in a confined work environment. Unplanned

removal of shields due to unexpected failures that require

complete removal of the shields or changing out of shield

components before the completion of the panel are that much

more difficult and hazardous. The safety issues and ground

control issues are interrelated. Poor shield performance can

lead to ground control problems, and stoppage in face advancement to remove and/or change out shield components can cause

the ground conditions to worsen and create a "snowball" effect.

Thus, proper shield design is essential to any longwall

operator. Engineering issues pertaining to shield design are

addressed by the author in another paper in these proceedings

[Barczak 2000a]. This paper provides several key points

pertaining to the procurement process, including the reasons for

buying a higher capacity shield, performance testing strategies

to ensure that the shield design is adequately evaluated, and the

benefits of buying extra shields and implementing a wellplanned preventive maintenance program once the shields are

purchased. In addition, traditional strategies of testing a shield

through a limited number of cycles are challenged with recommendations to test a shield to failure. The paper concludes with

proposed ideas and challenges for future shield designs.

EVALUATE YOUR CURRENT DESIGN FIRST

Before new shields are purchased, your current shield design

should be evaluated. Loading histories, if kept, can provide

valuable information to properly size your next support system.

Questions to answer include: How often were the supports at

yield pressure and when did this occur? Were there any ground

control problems, and what were they? Were the shields

rebuilt? Were any structural modifications made to the shields?

To what degree were the pin bores worn, and when did the wear

occur? What was the reliability of the leg cylinders? How

often did leakages occur? How often were the cylinders

rebuilt? How reliable was the control system for operating the

shields? Were there any warranty issues?

Performance testing of an aging shield can provide insight

into how much life expectancy is left in your current shields, as

well as reveal valuable information about making

improvements 

in the design of the new shields when they are purchased. The

National Institute for Occupational Safety and Health (NIOSH)

has developed a series of tests using the unique Mine Roof

Simulator load frame (figure 1) for conducting safety performance testing of aging longwall shields. These procedures are

described by the author in another paper in these proceedings

[Barczak 2000b]. The Mine Roof Simulator is an active load

frame that can apply both vertical and horizontal loading to a

shield and accurately simulate the in-service loading

conditions.2

 2

Further information on using the Mine Roof Simulator for shield testing

may be obtained by contacting the author at (412) 386-6557 or by e-mail at:

thb0@cdc.gov
247

Figure 1.—The Mine Roof Simulator at NIOSH’s Pittsburgh Research Laboratory.

KEY POINTS TO CONSIDER WHEN BUYING NEW SHIELDS

1. Identify your goals in buying a bigger shield.

Shield manufacturers now provide two-leg shields with

capacities approaching 1,200 tons, nearly four times the

capacity of the first shields installed in the United States in

1975 [Fiscor 1999]. Likewise, shield life expectancies are now

60,000 to 70,000 cycles, more than double what it was just a

decade ago. Although there is a natural tendency to buy the

biggest shield available, you should really consider your goals

when making this decision. Are you looking for more support

capacity to improve ground control, or is your primary goal to

extend the life of the shield, or do you wish to achieve both

goals?

If your primary goal is to extend shield life and your

previous support of lower capacity was providing adequate

ground control, then consideration should be given to derating

the higher capacity shield if you choose to purchase one.

Derating can be accomplished in two ways: (1) reducing the

setting pressure and (2) lowering the yield pressure. The goal

is to reduce the total loading on the support as a percentage of

the yield load rating of the support. The total shield load will

be 

the sum of the setting load and load developed in response to

the convergence. This total load will be limited by the yield

pressure of the leg cylinders. The load developed in response

to the convergence will be controlled primarily by the stiffness

of the leg cylinders. Since the stiffness increases with

increasing leg cylinder diameter, a higher capacity shield will

develop more load than a lower capacity shield for the same

convergence. Thus, unless the increased setting force results in

decreased face convergence, the increased setting force will

unnecessarily increase the overall shield loading.
248

Derating a shield by reducing the operating yield pressure

will control the total loading and associated component stress

development. This will provide an additional margin of safety

and extend the life of shields where high component stresses are

observed. Whether yielding is detrimental or not probably

depends on the rock mechanics and source of the shield

loading. If the convergence is caused by main roof weighting

that is irresistible in terms of the available shield capacity, then

promoting yielding at a slightly lower yield pressure is not

likely to have any significant impact on roof stability.

If you need the extra capacity for ground control and you

want to improve the life expectancy, then be prepared to pay a

premium for the shield and insist on a conservative design. A

conservative design can be qualified in terms of the component

stress developments, recognizing that the chance for failure

increases as the component stresses increase and that fatigue

and corrosion failures can occur for nominal stress developments well below the yield strength of the steel. Quantifying

stress limitations can be difficult since there are several factors

involved. However, a reasonable rule of thumb is that no

component should be stressed beyond 80% of the yield strength

of the steel under any in-service load condition, and stresses

should be below 60% of the yield strength for typical load

conditions. Classical structural design often requires that

stresses be kept below 50% of the yield strength to prevent the

occurrence of fatigue-related failures [Barczak 1999].

In summary, there is a common misconception that higher

capacity shields will be loaded to a lesser degree than lower

capacity shields simply because they have a higher support

capacity. Since the setting pressures have remained constant

and the stiffness increases as the capacity of the shield increases

because of the larger leg cylinder diameter, higher capacity

shields are just as likely to be fully loaded as lower capacity

shields under the current operating practices.

2. Make sure that performance testing is completed before

production shield fabrication begins.

A prototype shield should be fabricated and thoroughly

performance tested before fabrication of the production shields

begins. A shield consists of five major components (canopy,

base, caving shield, lemniscate links, and leg cylinders) all

connected together. Each component must be properly

designed to effectively transfer roof loading through the support

structure. If one component is underdesigned, additional load

may be transferred to other shield components, potentially

creating a domino effect that will likely reduce the life

expectancy of the support. Also, once a fabrication is made,

modifying the existing fabrication to correct the fundamental

design deficiency will generally not be possible or at best be

less effective than redesigning the component completely.

Additionally, modifying an existing fabrication can be difficult

when complex geometries and high-strength steels are involved.

Furthermore, the weight of a modern-day shield needs to be

minimized. Correcting a design deficiency by adding a

reinforcement plate to a fabrication generally will not result in

an optimum strength-to-weight ratio for that component. Once

the production shields are fabricated, any change in design can

be costly, and the tendency will be to minimize the cost by

avoiding the modification if possible or implementing the

lowest cost modification, which may allow the support to pass

the performance testing, but will not provide the operator with

the best design possible.

3. Make sure that performance testing is properly done.

Performance testing must properly simulate the in-service

load conditions. NIOSH has developed a set of safety

performance testing protocols that take advantage of an active

load frame called the Mine Roof Simulator. This biaxial load

frame can apply both vertical (compression) and horizontal

(racking) loads simultaneously to simulate the ground

movements and shield interaction. These test procedures are

described in detail by the author in another paper [Barczak

2000b] and provide the most direct simulation of the in-service

loading conditions [Barczak 1999]. In comparison, tests

conducted in a static load frame rely on external pressurization

of the leg cylinder to generate support loading. The limitation

imposed by the static load frame is the inability to directly

simulate the forward translation of the canopy due to slippage

of the support on the mine roof or floor and the face-to-waste

racking of the shield by the strata caving into the gob. These

conditions can increase the load development in the caving

shield and lemniscate links by several hundred percent.

Slippage of the canopy on the mine roof due to loss of

frictional resistance at the roof interface can be simulated in

static and uniaxial active load frames by placing rollers on the

canopy to create a "frictionless canopy test." Unfortunately, the

zero-friction test with rollers on the canopy is not very practical

for extensive cyclic loading since the rollers may require

frequent adjustments due to frictional effects. Thus, this test is

typically not included in the manufacturer's performance testing

protocol. If the frictionless canopy test cannot be conducted,

then the support should be configured so that it is standing on

the toe of the base; this will require horizontal constraints at the

toe of the base and rear of the canopy (at the caving shield

connection). The horizontal force couple generated by this

configuration allows load transfer to the caving shieldlemniscate assembly, similar (although not as severe) to that

obtained with the zero-friction test.

Likewise, face-to-waste racking of the shield cannot be

directly simulated by a static or uniaxial active load frame. The

contact configuration that most closely duplicates the caving

shield-lemniscate assembly response for this condition in static

or uniaxial load frames is to configure the support so that it

is simply supported on the rear of the base. Face-to-waste

racking of the shield canopy causes a reversal in the state of

stress in the lemniscate links from compression to tension.
249

Because reversal of the loading accelerates fatigue failures, both

the base-on-toe and the base-on-rear configuration should be

incorporated into safety testing protocols conducted in static or

uniaxial load frames.

Longwall operators in cooperation with the shield manufacturers have developed and refined performance testing

procedures for longwall shields using static load frames.

Consolidation Coal Co. in particular has led this effort. The

so-called "Consol Test" consists of several combinations of

canopy and base contact configurations that are categorized as

follows:

(1) Offset yield loading;

(2) Base toe loading;

(3) Three-point canopy torsional loading;

(4) Two-point canopy torsional loading;

(5) Side shield loading and base torsional loading;

(6) Diagonal base contact;

(7) Three-point base contact torsional loading;

(8) Symmetric base edge loading;

(9) Asymmetric base edge loading;

(10) Leg socket loading; and

(11) Canopy dishing.

These tests are designed to be used in static- or single-axis load

frames. These test configurations have proven to be among the

most effective protocols for performance testing of shield

supports in a static frame. The operator should consult the

support manufacturer and discuss these or equivalent tests to be

used for the performance testing program.

4. Measure the load development as part of the performance

testing.

It is not enough to simply measure the leg pressures during

performance testing. Some measurement of load transfer

through all the support components should be conducted as part

of the test program to verify the effectiveness of the test

procedures and to evaluate the stress developments within the

shield. This can be done with a few strain gauges to measure

nominal load (stress) development in each component.

Photoelastic plastic can be used to isolate areas of expected

stress concentration.

5. Consider testing a shield to failure.

Most companies conduct performance testing only through

a limited number of cycles, generally equal to the warranty

period provided by the shield manufacturer, which is typically

50% to 60% of the life expectancy of the shield. This practice

is adequate to discover fundamental flaws in the shield design.

However, for a well-designed shield, this approach will not

evaluate fatigue failures that occur near the end of the shield

life. In addition to providing a more definitive estimate of the

life expectancy of the shield, testing a shield to failure will

provide insight into the nature and severity of the fatiguerelated problems that will eventually occur. It will also provide

the mine operator with insight as to when to look for these

problems because they are often difficult to see when they first

develop, but can be catastrophic if left unattended. In addition,

by testing the shield to failure, the impact of the failure(s) on

the shield structural integrity and performance of the shield can

be determined beforehand.

In summary, by testing the shield to failure, the mine

operator will be in a much better position to (1) plan for the

next shield procurement cycle, (2) develop a monitoring plan

for inspecting the shields for failures, and (3) develop a strategy

for repair and modification of the shields when failures do

occur. The manufacturers, and in turn the mine operators, will

also benefit from failure testing by providing critical

information on the limitations of current shield design that can

lead to design improvements in the future.

6. Consider buying several extra shields.

There is clear evidence that mines that purchase additional

shields and institute good preventive maintenance program will

maximize the useful life of their longwall shields. A good rule

of thumb is to purchase 10% more shields than is necessary for

the face. For example, a 1,000-ft face would require

174 shields (1.75 m wide). If 17 extra shields were purchased

so that 17 shields could be removed from operation during each

longwall move, the entire face would be recycled in 10 panels

of mining. Mining a 10,000-ft-long panel will require about

3,300 shield cycles; thus, each shield will be changed out at

least once during the shield's warranty period. More

importantly, once the warranty period is over and the shields

have been in operation for 35,000-40,000 cycles, a detailed

inspection and rebuild program can be instituted. Problems

such as leg cylinder rebuild and restoring pin tolerances to

design specifications could be corrected as they arise, and

preventive measures, such as painting and crack inspection and

repair, could be done on a regular basis. Although this adds

cost to the initial purchase, a good preventive maintenance

program will pay big dividends in maximizing the life of a

shield.
250

CONTINUE PUSHING THE ENVELOPE

Improvements in shield design have been made largely

through mine operators pushing the envelope in terms of

expectations. Ten years ago, a life expectancy of 60,000

loading cycles was unheard of, yet today it is the standard.

Likewise, 1,200 tons of support in a two-leg shield was

unthinkable 20 years ago, yet shield capacities continue to

increase as each new generation of supports is developed.

Although there certainly are limitations in any engineering

design and factors such as the weight of the shield will soon

pose a barrier to increased size with current technologies, one

must remember that the unthinkable 20 years ago is now the

standard. Additional improvements in shield design will be

made as long as mine operators continue to push the envelope

in shield design.

Below are some items to consider for setting goals for the

next generation of shield supports.

1. 100,000 Life Cycle Expectancy.CThe increases in life

expectancy realized during the past 10-20 years should continue

in the near future. It is not inconceivable for a shield to survive

100,000 loading cycles. To reach this goal, some

improvements in design may be necessary or at least closer

attention will need to be paid to fundamental design practices.

In particular, components will need to be sized such that stress

levels are kept at moderate levels with respect to the yield

strength of the steel. Stress concentrations due to sharp changes

in geometry must be avoided, and weldments must be of high

quality. More corrosion-resistant materials may be used in

certain areas to reduce stress corrosion. Make sure that the link

bores have sufficient bearing areas to avoid high levels of

stress, and allow for the additional stresses caused by corrosion

in the design of the link bores.

2. Hydraulic Diagnostic Capability.CAll shields will

experience leakage in the leg cylinders or other hydraulic

components that degrade the support capability. An algorithm

should be specifically written into the control computer that

monitors leg pressure histories that will provide the longwall

coordinator with a record of bad leg cylinders that need to be

repaired. Another option is to have the computer reset leg

cylinders that are leaking below a designated setting pressure.

While this capability exists on some modern faces, it should

become a standard part of the operating system.

3. Smart Load Control.CThe benefits of derating a shield by

reducing the total loading on the shield were discussed earlier

in this paper. A challenge for future shield designs would be

for the control computer to optimize the set pressure by

monitoring load development history and adjusting the set

pressure to minimize overall shield loading once a ground

reaction curve for the longwall face is established. The set

pressure could be adjusted as the conditions change to

continually optimize the support capacity utilization.

4. Lubricated Link Joints.CPin joints are necessary to

accommodate the kinematics of a shield, but these joints are by

far the leading cause of structural rebuild. Efforts should

continue to enhance the use of wear-resistant materials, such as

impregnating the pins with zinc phosphate, but consideration

should also be given to a lubricated joint to reduce the wear.

5. Composite Material Design.CEven with the use of highstrength steels, modern shields weigh as much as 30 tons.

Weight is a major barrier in increasing the shield widths beyond

2 m with the current steel constructions. Composite materials

of equivalent strength weigh only one-fourth that of steel

fabrications. Although several engineering and economic

issues need to be explored before the feasibility of using hightechnology composite materials for shield construction can be

determined, this could lead to a new generation of shield

supports in the future.

6. Forecasting Periodic Weighting and Heavy Roof

Loading.CMost modern shield systems are capable of capturing

shield loading histories to the point where we are overloaded

with data. These data, if properly managed, can be useful in

evaluating leg cylinder leakage and optimizing setting

pressures, as suggested in items 2 and 3 above. Another area of

value is the prediction of periodic weighting intervals. There

has been some research in this area already. For example, NSA

Engineering, Inc., developed a shield monitoring program that

couples with its GeoGuardJ software [Sanford et al. 1999].

The system was recently tested in an Australian mine where

severe face weighting was observed, resulting in hazardous face

conditions and weeks of lost production.

7. Advanced Component Load Measurement.CCurrently,

only the leg cylinder pressure is monitored. No loading

information is obtained on any other components. It would not

be difficult in the design process to include strain gauges on

selected components that could be monitored by the

computerized data acquisition system. This need not be done

on every production shield; in fact, two or three shields could

serve as instrumented shields and provide valuable information

on the actual load conditions observed underground. This

information could then be used to refine performance testing, as

well as help to identify load conditions that actually contribute

or cause structural failures. Even measurement of the link

loading alone would provide valuable information that is

currently not available.
251

8. Constant Set Leg Cylinder Design.—Current shields use

a two-stage cylinder design where the first stage extends and

retracts first. This operation typically causes the first stage to

be near full extension most of the time. In addition, seal leakage, which occurs on almost all aging shields, causes the bottom

stage to extend outward and further promotes full extension of

the bottom stage. Since the setting force and the subsequent capacity of the shield to resist roof loading is reduced in proportion to area differential between the top and bottom stage

when the bottom stage is fully extended, this is not a good

design to ensure that the full capacity of the shield is used. In

other words, why buy a 1,000-ton shield when its performance

is degraded to that of a 500-ton shield most of the time? There

are alternative designs available, most notably that employed by

Fazos, Inc., which do not use the conventional two-stage design

used in most modern shields. The Fazos design uses a central

core to eliminate the use of independently acting stages. Although this design works, it has not been adopted by the major

shield manufacturers. An alternative approach would be to configure the standard two-stage design with a system that would

cause the top stage to extend first. This could be done by using

a nominal hydraulic pressure in the retract annulus of the

bottom stage during the extend operation. For current shields,

a mechanical device (strap or chain holster) could be built that

would attach to the cylinder casing and the base that would

prevent the bottom stage from extending. This would not be

used on every mining cycle, but would be done periodically as

part of a preventive maintenance program to restore full setting

capability to the leg cylinders, which are routinely at full

bottom-stage extension.

CONCLUSIONS

Improvement in shield design continues to be made with

greater life expectancies than ever before. This can certainly

benefit mine operators, but the increased cost of modern shields

also places greater emphasis on ensuring that the shield design

is good since the consequences a poor design can be

catastrophic from both an economic and a mining perspective.

Proper planning is essential for a new shield procurement.

Sufficient time must be built into the procurement process to

ensure that a prototype shield is adequately tested before

fabrication of the production shields begins. Otherwise, design

changes may not be able to be made in time and/or the

modifications will not be as effective as changing the original

design. Performance testing needs to be done properly, and

every effort should be made to simulate the in-service

conditions that occur in the mine. Testing a shield to failure,

even if it incurs additional costs, will give insight into when and

what to expect when fatigue-type failures plague the support as

it ages and needs to be rebuilt or nears the end of its useful life.

Purchasing additional shields and implementing a preventive

maintenance plan that allows for scheduled rebuild of hydraulic

cylinders and inspection and repair of structural components at

the onset of problems will pay big dividends in extending the

life of the shields.

One misconception is that modern shields by virtue of being

higher capacity will last longer than lower capacity

shields of previous generations. Modern-day shields are just as

likely as lower capacity shields of previous generations to be

fully loaded. In fact, the higher strength steels used in modernday shields generally are prone to more brittle and catastrophic

failures when excessive loading occurs than previous

generation shields. This fact, coupled with the higher cost and

longer life expectancy, makes it more important than ever that

shields be properly designed. Thus, it is important to determine

the design quality and limitations of the shield during the

procurement process.

Improvements in shield design will continue as long as

operators continue to demand improvements to which there are

engineering solutions. A life expectancy of 100,000 cycles is

a reasonable goal that can be accomplished with current

technology. Modern longwall faces all have computers collecting data and providing automation of the support operation.

The next logical step is to use the data to further improve the

diagnostic capabilities and optimize the shield performance.

This will be a challenge to shield operators and will require

some experimentation on operating longwall faces, but again is

within the realm of current technology. The next major improvement in shield design may be the use of composite

materials to lighten the weight of the shield. Reducing weight

could break the 2-m width barrier that limits current designs

using high-strength steel fabrications.

REFERENCES

Barczak TM [1999]. Modern shield technology: better than ever, but still

not perfect. In: Peng SS, Mark C, eds. Proceedings of the 18th International

Conference on Ground Control in Mining. Morgantown, WV: West Virginia

University, pp. 149-163.

Barczak TM [2000a]. Examining longwall shield failures from an

engineering design and operational perspective. In: New Technology for Coal

Mine Roof Support. Pittsburgh, PA: U.S. Department of Health and Human

Services, Public Health Service, Centers for Disease Control and Prevention,

National Institute for Occupational Safety and Health, DHHS (NIOSH)

Publication No. 2000-151, IC 9453.

Barczak TM [2000b]. NIOSH safety performance testing protocols for

standing roof supports and longwall shields. In: New Technology for Coal
252

Mine Roof Support. Pittsburgh, PA: U.S. Department of Health and Human

Services, Public Health Service, Centers for Disease Control and Prevention,

National Institute for Occupational Safety and Health, DHHS (NIOSH)

Publication No. 2000-151, IC 9453.

Fiscor S [1999]. U.S. longwall census '99. Coal Age Feb:30-35.

Sanford J, Mahoney S, Conover DP, DeMarco MJ [1999]. Shield

monitoring to forecast severe face weightings at the South Bulga Colliery,

NSW, Australia. In: Peng SS, Mark C, eds. Proceedings of the 18th

International Conference on Ground Control in Mining. Morgantown, WV:

University of West Virginia, pp. 164-176.
253

INSTRUMENTS FOR MONITORING STABILITY 

OF UNDERGROUND OPENINGS

By Mark K. Larson,1 Douglas R. Tesarik,2 J. Brad Seymour,1 and Richard L. Rains3

ABSTRACT

For several years, researchers from the Spokane Research Laboratory of the National Institute for

Occupational Safety and Health (NIOSH) have been using geotechnical instruments in underground mines to

study ground control problems and develop means of reducing accidents and fatalities caused by ground falls.

This paper describes many of the different types of instruments, sensors, and data acquisition equipment that

have been used for these studies; briefly explains the advantages and disadvantages of various sensor

technologies; provides practical recommendations regarding the use of specific instruments and data

acquisition systems; and outlines a general approach to the design and implementation of a successful instrumentation plan.

A wide variety of instruments are commercially available for measuring deformation, strain, stress, and/or

load. If used correctly, these instruments can provide important quantitative information regarding the mininginduced behavior of the host rock, the performance of ground support systems, and the safety and stability of

underground workings. Data collected from these instruments can warn mine staff of impending ground

control failures or hazardous working conditions, as well as provide valuable information for the design of

ground support systems and the configuration and sequencing of mining excavations. We hope that the

practical information presented in this paper will encourage a more widespread use of geotechnical instruments

in underground mines and lead to reliable measurements that aid the engineering decisions affecting the safety

of miners.

1

Mining engineer.

2

Mechanical engineer.

3

Geologist.

Spokane Research Laboratory, National Institute for Occupational Safety and Health, Spokane, WA.
254

INTRODUCTION

Researchers at the Spokane Research Laboratory (SRL) of

the National Institute for Occupational Safety and Health

(NIOSH) have been placing instruments in mines for several

years to study means of reducing accidents and fatalities caused

by ground falls. This paper describes ways of using these

instruments. The goal of this paper is to share information and

tips that encourage mining operators to use instruments to aid

in decisions affecting the safety of miners.

In today's world of mining, layout and roof support are

designed so that pillar and entry safety factors are only a little

above the critical point of failure. Under varying conditions,

the amount and type of roof support may greatly affect stability.

It is therefore desirable to optimize roof support, i.e., the

density and/or type should be sufficient to minimize the risk of

ground fall, yet minimize costs. Technical personnel must

determine this point using good engineering judgment.

Numerical models can be valuable tools in providing answers to "what if" questions, but they may provide poor information if erroneous assumptions or poor input values are used.

For example, a homogeneous continuum model may not be

sufficient if joints play a significant role in roof falls. Verification and tuning of models is a step that is quickly forgotten in

the rush for an answer.

Instruments have been used increasingly in mines to measure

deformation, stress, strain, and load. Such measurements serve

two purposes. First, they provide quantitative information that

can be useful in determining the mechanics of stability and in

aiding engineering decisions. Second, they can be useful in

verifying and tuning numerical models.

Bauer [1985] briefly described many types of available

ground control instruments. Hanna [1985] comprehensively

described instruments and measurement theory as applied to

foundation engineering and construction. Dunnicliff [1988]

updated information on instruments and how to plan monitoring

programs. While new instruments have been designed and

improvements made in older ones, it is not the purpose of this

paper to give a comprehensive update. Rather, the purpose of

this paper is to outline an approach to instrumentation that we

have used for several years, describe the basic types of sensor

technology, describe our experience with various instruments,

list manufacturers of these instruments known to the authors,

and provide pointers on continuous monitoring with a data

acquisition system.

INSTRUMENTATION PLANNING

Design and implementation of a successful instrumentation

plan requires careful consideration of the following factors:

$ A fundamental understanding of the overall rock

mechanics, geology, and general site conditions;

$ Identification of the critical mechanisms or geotechnical

parameters that need to be measured;

$ Clearly defined objectives for each of the measurements

to be taken;

$ Estimates of the magnitude or anticipated range for each

of these measurements;

$ Selection of instruments, transducers, and data acquisition

equipment appropriate for the desired measurements and site

conditions;

$ Availability of experienced personnel to install, troubleshoot, and monitor the instruments; and

$ Established procedures for data collection, analysis, and

reporting.

In general, instruments should not be installed until appropriate

provisions have been made for collecting, analyzing, and

reporting the instrument data.

Programmable data acquisition systems with multiplexers

provide the user with the opportunity to scan and record

measurements frequently from a large number of instruments.

To ensure that the maximum benefit is derived from instrumentation programs, especially those capable of collecting

enormous amounts of data, careful planning is necessary.

Entire chapters in geotechnical instrumentation books are

devoted to the subject of planning instrumentation programs

and interpreting recorded data [Dunnicliff 1988; Hanna 1985].

Similarly, the number of papers presented at instrumentation

conferences that address planning issues is sometimes large

enough to comprise a technical session [Leung et al. 1999].

A successful instrumentation program starts by defining the

purpose of installing the instruments and ends with implementing the decisions derived from interpretations of the data.

Dunnicliff [1988, 1999] offers a 21-step checklist in which each

step in a successful instrumentation program is described.

Because comprehensive information about instrumentation

planning is available in geotechnical literature, the following

discussion focuses specifically on experience gained from

installing instruments and data acquisition systems in underground hard-rock and coal mines, maintaining these systems,

reducing the collected data, and interpreting the data for

immediate and future use.

Because one of the goals of an instrumentation program is to

monitor ground stability and take corrective action for any
255

∆E ' Vx

R1R2

R1%R2

2

∆R1

R1

& ∆R2

R2

% ∆R3

R3

& ∆R4

R4

(1)

imminent problem, instruments that sense displacement and

stress changes are commonly used. Instruments should be

checked to ensure that they have not been damaged during

shipping and handling. In general, displacement transducers

should be recalibrated after they have been received, especially

if the sensor is a potentiometer or a Wheatstone bridge circuit

and if cable is added between the sensor and monitoring device.

This action is helpful because the new calibrations will account

for resistance caused by the cable. For vibrating-wire transducers, factory calibrations are generally sufficient because calibration constants are not changed by adding cable.

However, operation of the transducers should be verified

before they are used underground. It is useful to have calibration sheets for displacement transducers at installation sites to

assist in positioning transducers in the desired initial positions.

It is important to place the instruments at locations where the

expected parameter changes can be measured accurately. For

tabular ore bodies, estimates of change in values, such as stress

or strain in pillars, abutments, and backfill, can often be

calculated using the tributary load method. More complicated

mine geometries may require numerical models to identify

areas in the mine that will undergo high stress changes. For

extremely critical areas near shafts or shops, it may be

necessary to calculate changes in stress and displacement at

specific points using a numerical model before installing

instruments. This information can then be used to select the

appropriate instrument range. If a two-dimensional numerical

model will be used with instrument data collected during

mining to evaluate mine stability, it is important to place an

adequate number of instruments along the modeled cross

section to provide calibration points for the model.

Installing two instruments that measure different parameters

at the same location increases the probability that collected data

will be useful. For example, pillar stability could be assessed

with a stress-change measuring device such as a biaxial

stressmeter and a displacement measuring instrument such as a

horizontal borehole extensometer [Seymour et al. 1998]. Stress

changes in the pillar can be verified by a borehole extensometer, which measures pillar dilation. Similarly, stress

changes in backfill could be measured by embedment strain

gauges and cross-checked with earth pressure cells [Tesarik et

al. 1995]. Borehole extensometers in the mine roof may help

analyze loading mechanisms [Tesarik et al. 1999]. Clustering

instruments in this manner was useful in numerous instrumentation programs and served to verify measurements and aid

in data interpretation. Clustering also provided backup data

when one of the instruments or its cable was destroyed or

malfunctioned.

SENSORS AND MONITORING

The authors have primarily used the following sensor

technologies for instruments installed in underground mines:

strain gauges configured in Wheatstone bridge circuits, potentiometers, and vibrating-wire transducers. These sensors can be

read either manually with a portable electronic readout box or

at programmed intervals using a computerized data acquisition

system set up in a central location near the instruments. Manual

data collection methods are typically used in situations where

only a limited number of instruments are monitored; individual

instruments are installed in remote, isolated locations; it is

impractical to route the instrument cables to a central location;

personnel experienced with the use of dataloggers are not

available; or the additional expense of a data acquisition system

is prohibitive. Although manual readings can provide reliable

information about the stability of underground openings

[Tesarik et al. 1991, 1999], these readings are usually taken too

infrequently to allow a meaningful analysis of measurement

quality and are subject to human error. If possible, instruments

should be monitored with a data acquisition system, and manual

readings should be taken periodically with an electronic readout

box to verify the datalogger measurements. Frequent readings

not only permit better judgment about the quality of the

measurements, but also help associate trends in the instrument

data with discrete mining events. The use of dataloggers and

their ancillary components is described below in the "Data

Acquisition System Basics" section, along with several

recommendations for improving the likelihood of quality

measurements.

SENSOR TECHNOLOGIES

Wheatstone Bridge

A Wheatstone bridge is a circuit (figure 1) that provides a

signal output directly related to strain. If R1, R2, R3, and R4 are

the respective resistances of arms 1 through 4 in figure 1, and

R1R3 ' R2R4, then the differential output voltage, )E, may be

represented by
256

Figure 1.—Wheatstone bridge wiring diagram.

 Figure 2.—Quarter-bridge wiring diagram. A, Two-wire

leads; B, three-wire leads.

if second-order terms are neglected [Dally and Riley 1991]. In

equation 1, Vx is the applied excitation voltage and )R1, )R2,

)R3, and )R4 are the changes in resistance of strain gauges 1,

2, 3, and 4, respectively. If all arms of the bridge are active (or

all are subject to strain-induced resistance change), then the

circuit is called a full Wheatstone bridge. A half-bridge is

described when R3 and R4 are fixed and R1 and R2 vary with

strain. A quarter-bridge is described when only R1 varies with

strain and the other arms of the bridge are fixed.

Load cells and pressure transducers are examples of typical

full-bridge sensors. An instrumented bolt or cable has strain

gauges positioned at various distances along its length on

opposite sides of the bolt. Each strain gauge is read by

completing the bridge with fixed precision resistors (quarterbridge circuit), which are usually located near the data

acquisition unit. It is possible to use one set of completion

resistors (R2, R3, and R4) by using switches on the strain gauge

leads. The switches to a particular strain gauge (R1) are closed

while the switches to the other strain gauges are left open. The

next gauge may be read after opening the closed switches to the

current strain gauge and closing the switches to the next strain

gauge. The completion resistors should be kept as close to the

temperature of the active gauge(s) as possible to avoid

thermovoltaic errors.

In the case of a quarter-bridge circuit, long wire leads may

affect the circuit by reducing sensitivity (see figure 2A, where

R1 ' Rg % 2RL). This sensitivity loss is typically compensated

for by a reduction in the gauge factor used to convert voltage

signal to strain. The loss resulting from signal attenuation may

be cut approximately in half by using the three-wire quarterbridge circuit, as shown in figure 2B.

Advantages of Wheatstone bridge technology are that

response is immediate and very sensitive and the instrument can

be used for measuring static or dynamic strain. Two disadvantages are that the datum-state reading shifts over time

(zero drift) and that the instrument may respond secondarily to

temperature. Much of the drift may be attributed to glue creep.

Another disadvantage of the Wheatstone bridge is the low

level of the signal, )E. In a mine environment, there are

several possible sources of external current or noise that may

overshadow or drown the signal, including high-voltage power

cables, leaky-feeder-type communication systems, machine

noise, and multiple grounds. In most cases, such effects from

external sources are not present. Instrument cables should be

hung away from other power cables. Power cables should be

crossed only when necessary. A shield should always be placed

around the power cable, and the instrument cables should be

routed so that they cross the power cable at a right angle.

Multiple grounds have periodically been a problem,

particularly with instrumented bolts, where a connection to an

earth ground may run through the bolt. If another earth ground

has already been established (for example, near the datalogger),

a ground loop is formed, which may drive bridge signals up or

down, often completely out of its expected range. For this

reason, the authors now choose to not connect the datalogger's

power supply ground to an earth ground.

Potentiometer

Figure 3 shows a typical potentiometer circuit used to sense

displacement. A resistive element is excited with a constant,

known voltage Vx. A wiper contacts the resistive element along

its length. One-dimensional movement of the wiper with

respect to the resistive element is sensed by measuring the

voltage of the wiper with respect to the ground. The measured

wiper voltage ranges from 0 to the excitation voltage,

depending on its contact location on the resistive element.
257

Figure 3.—Potentiometer circuit.

Potentiometers may be manufactured to cover a variety of

displacement ranges. Sensitivity depends on the excitation

voltage and the voltage monitoring device.

The authors have not encountered noise problems with

potentiometers because the level of signal used has typically

been 2 to 2.5 orders of magnitude above that of the Wheatstone

bridge circuit. However, the user should be aware of the

potential to burn up the wiper-resistive element contact if a

large electrical current is discharged through the wiper

following measurement. One precautionary measure is to

connect a resistor to the wiper to restrict the amount of current

that can pass through it. The authors have used a 1-kS resistor

on the wiper with a 20-kS resistive element.

In the case of long lead wires, the user can choose to use

the circuit as is. However, the circuit must be calibrated with

the long lead wires before installation. The authors have

determined applied excitation voltage using a fourth wire

(dashed line in figure 3) to measure the voltage drop across one

lead length. The voltage applied at the resistive element is the

excitation voltage applied at the datalogger less the drop across

two lead lengths. The voltage drop across one lead length is

determined with the electric potential between the dashed wire

and analog ground. By assuming equal voltage drop in both

lead wires, the applied excitation at the resistive element is

determined using the following equation:

Applied excitation at the resistive element . Vx & 2(Vx & EL), (2)

where Vx ' excitation voltage applied at the datalogger;

 and

EL ' the measurement in volts between the dashed

 lead and analog ground.

The potentiometer reading may be normalized with this applied excitation voltage. The raw reading, applied voltage, and

normalized reading should be stored in the measurement data

file.

Vibrating-Wire Transducer

Vibrating-wire transducers are used in many instruments,

including load cells, deformation gauges, surface and embedment strain gauges, earth pressure cells, pressure sensors for

piezometers, and liquid level settlement gauges [Dunnicliff

1988; McRae and Simmonds 1991; Choquet et al. 1999]. A

steel wire is tensioned between two clamps so that it can vibrate

at resonant frequency. As the clamps are physically moved

relative to each other by applied force to the sensor (displacement transducers use a linear spring to convert displacement to a proportional force on the vibrating wire), the

resonant frequency changes. A coil is positioned near the wire.

When current is run through the coil, the resulting magnetic

field in effect plucks the wire, causing it to vibrate at a resonant

frequency reflective of the external force. The vibrations, in

turn, cause voltage fluctuations in the coil that correspond to the

vibrations. A handheld readout box or automated data

acquisition system measures the time it takes to complete a

given number of vibration cycles and computes the frequency.

Sometimes an interface unit is used to boost the excitation

voltage. Signal conditioning and conversion to digits of frequency squared or engineering units are performed by a

datalogger.

In the case of a continuous readout, two coils are used. One

coil electronically plucks the wire and senses the vibrationcaused voltage fluctuations. This frequency is calculated. The

second coil vibrates the wire at the same frequency. As

frequency changes, so does the plucking frequency of the

second coil.

There are several distinct advantages to using vibrating-wire

transducers. First, the readings are not sensitive to long lead

wire lengths because frequency, rather than voltage level, is

being measured. Low-capacitance cable may be required for

lead lengths greater than 200 m (650 ft). This means that a

cable can be cut in the field to the length needed without

offsetting the output signal. Moreover, Choquet et al. [1999]

report that manufacturers have perfected the manufacturing

process to the degree that there is very little drift in the reading

at no load or zero displacement over time and only small

changes occur in calibration slopes (within ±0.5%) after nearly

4 years. McRae and Simmonds [1991] report that all of their

tests, some lasting for over 10 years, show that if wire stress is

kept below 30% of yield, long-term drift will be minimal. For

pressure transducers where minimum stress is only 13% of

yield, long-term drift is insignificant. McRae [2000] reports

that his tests show virtually no drift after 15 years. This

performance makes vibrating-wire technology very desirable

for long-term monitoring.

One problem that has surfaced from time to time is that

the datalogger will fail to obtain a meaningful reading when

it is monitoring a vibrating-wire transducer that incorporates

a vibrating-wire interface to the datalogger. One explanation
258

Figure 4.—Campbell Scientific CR10QM wiring panel.

may be that the selected frequency sweep range did not include

the frequency to be read. More often, however, it is a contact

problem where wires are clamped at terminals. These contacts

must be tight and of good quality. In the authors' experience,

when a poor connection was discovered during sensor checks

or calibrations, improving the contact generally solved the

problem. A vibrating-wire readout box provides a good

backup, as it provides more signal filtering. The authors have

never encountered a problem when using a vibrating-wire

readout box instead of a datalogger. Experience has helped

minimize problems encountered when the sensors are

monitored with a data acquisition system.

DATA ACQUISITION SYSTEM BASICS

The authors have used dataloggers from Campbell Scientific,

Inc., for several years. Models 21XQM (now discontinued by

the manufacturer) and CR10QM are approved by the Mine

Safety and Health Administration for use in methane-air

atmospheres. Although operation of one of these dataloggers

is relatively simple, learning to use the system may require a

little effort. There is a learning curve for the dataloggers and

related components, but unattended instrument monitoring and

recording are usually well worth the effort required to learn to

use the system. In this section, basics of using these

dataloggers and associated components are discussed to help

the user considering sensor monitoring with an unattended data

acquisition system.

The dataloggers collect two basic measurements: singleended voltage and differential voltage. The difference between

these types of measurements will be aided by reference to figure 4, which shows a CR10QM wiring panel with labeled

terminals.

A single-ended measurement is a measurement of voltage

with respect to ground. For example, a single-ended measurement would be measured between two wire leads. One lead

must go into one of the analog ground (AG) terminals. The

other lead must connect to one of the input channels (1H

(high), 1L (low), 2H, 2L, etc.). The input channel must be

specified with a measurement command in the datalogger

program. For example, a potentiometer is read with a singleended connection to the datalogger. A vibrating-wire sensor

with an in-line interface is connected to the CR10QM in the

same way; however, instead of a single-ended measurement, the

datalogger measures time for a certain number of vibration

cycles and converts that measurement to frequency squared or

engineering units.

A differential measurement is a measurement of voltage

in one lead with respect to that in another lead. The two leads

are connected to differential channel input terminals (1H and

1L, or 2H and 2L, or 3H and 3L, etc.). The voltage of the high

(H) terminal is measured with respect to the low (L) terminal.

For example, Wheatstone bridge circuits are measured by

connecting the signal leads to a set of differential channel input

terminals on the datalogger.

A CR10QM datalogger has 6 differential inputs or 12 singleend inputs. The 21XQM has 8 and 16 inputs, respectively. The

number of sensors monitored may be extended by using

multiplexers. A multiplexer is an electronic switching unit with

mechanical relays. It has 4 common terminals and 16 channels

of 4 terminals each. A multiplexer is activated by sending and

maintaining a voltage that is greater than 3.5 V from a datalogger control port (setting the port high) to the multiplexer's

"reset" terminal. When a pulse is sent from a control port of the

datalogger to the multiplexer's "clock" terminal, switches close

between the multiplexers' four common terminals and the four

respective terminals of channel set 1. The next pulse opens

these switches and closes the switches between the common

terminals and the four terminals of channel set 2. The multiplexer is deactivated by removing the current through the

control port (setting the port low).

Wires are run from datalogger terminals (inputs, excitation,

and ground) to the four common terminals on the multiplexer.

Sensor leads are connected to the appropriate channel set

terminals.
259

The multiplexer increases the number of possible sensors

that may be monitored through a datalogger input terminal or

set of terminals. For example, the number of vibrating-wire

transducers that may be monitored through a multiplexer is 32

(2 per channel set). Potentiometers require an excitation lead,

and Wheatstone bridges require excitation and ground leads.

Therefore, it is possible to monitor 16 potentiometers or 16

Wheatstone bridges through 1 multiplexer. Sensor types are

generally not mixed on the same multiplexer.

The addition of multiplexers to a datalogger adds some

complexity to reading sensors. When a clock pulse is sent to a

multiplexer, a set of mechanical relays is closed. To ensure that

contact has been made, the user must insert a programmed

delay before the sensor is read. Typically, a delay of 0.02 to

0.03 sec is sufficient. Also, a delay of 0.05 sec is good practice

follow-ing multiplexer deactivation.

PC208W software and interfaces are used for communications, viewing data, and programming. For example,

one subprogram provides a real-time connection to the datalogger from a personal computer through an optical interface.

Another subprogram is used to communicate with a storage

module through an RS-232 cable and interface. Another

application is used to view data, and another to process

information from those data.

A good programming structural sequence is toC

1. Read the sensors, usually in groups (in a loop) by

multiplexer (readings are stored in input locations);

2. Read sensors internal to the datalogger (such as coldjunction temperature, battery voltage, program signature, etc.);

3. Set the output flag;

4. Read time and date;

5. Set resolution (usually high) and write readings from

input locations to final storage; and

6. Send the data to an output device, such as a storage

module.

Before writing a datalogger program, all circuits should be

diagramed from the datalogger to the sensor so that it is clear

what programming instructions and parameters are to be used.

With this information, programming is relatively simple. For

the user's first program, it may be a good idea to contract with

Campbell Scientific for programming assistance. Once the user

understands that program, additional programming is easy.

Once the system is installed, one potential operating problem

involves electrical grounds. Until recently, it was the authors'

practice to connect all cable shields to shield terminals and

connect the terminals to one earth ground near the datalogger.

In addition, the authors would connect the datalogger power

supply ground to the same earth ground. While this practice

seems logical for protecting transducers and the datalogger

when a cable is compromised, it can also affect readings if the

system is connected to a Wheatstone bridge. Ground potentials

can vary significantly between sensor locations. In fact, one of

the authors has measured a difference in ground potential of

approximately 1 V between points in a coal mine roof just a few

feet apart. When the instrument itself becomes part of a

ground, as can happen with instrumented bolts and cables, any

difference in ground potential between the instrument and the

ground near the datalogger induces a ground loop and affects

measurements. In such cases, it is not a good practice to

connect shields at the multiplexers or datalogger. Recently, the

authors began leaving the datalogger ground floating (not

connected to an earth ground).

The authors always collect and store raw analog measurements. Sometimes the raw data are also processed and stored,

but raw data are always saved. This procedure provides a

backup in case of an inadvertent mistake in processing.

Lastly, the authors strongly recommend spending some time

in the laboratory performing checks on the instruments and

datalogger before installing the instruments in the field. This

check avoids many problems that may be more difficult to solve

underground or after installation. Monitoring sensors over a

period of time in the laboratory can identify many problems.

Similarly, instruments should also be checked at the installation

site before they are installed. Experience has shown that use of

connectors should be avoided if possible. The authors prefer to

hard-wire instrument cables to datalogger and multiplexer

terminals. This avoids problems such as cold solder junctions

and the abuse that connectors always suffer underground as

cables are run. In the event that use of connectors is unavoidable, cold solder junctions may be found by identifying

erratic sensor readings when the connector is heated with hot

air.

INSTRUMENTS

Experience is important to increase the quality and

probability of success in installing instruments. In the sections

that follow, specific instruments and instrument types that the

authors have used are described. Comments and tips are

offered.

DISPLACEMENT INSTRUMENTS

The authors have used both potentiometer (linear and rotary)

and vibrating-wire sensors in instruments that measure

displacement. Some instruments were purchased from various
260

Figure 5.—Instrumented bolt (after Signer and Lewis [1998]).

 Figure 6.—Instrumented cable bolt (from Martin et al.

[2000]).

manufacturers, while the hardware of some other instruments

was manufactured in-house. The type of instrument used

should depend on how much displacement is expected, number

of required in-borehole anchors, instrument accuracy, stability

over time, and operational constraints. For small expected

displacements over a long period of time, grouted anchors or

hydraulic anchors may be more desirable than spring anchors.

For larger expected displacements, a spring with attached piano

wire running to string potentiometers mounted at the borehole

collar may be sufficient.

One of the authors used rotary potentiometers with a geared

spindle that was supposed to mesh with a threaded rod.

However, the gears sometimes skipped because the gear teeth

and thread profile did not match well. The skipping problem

was also very dependent on the force between these members.

Force adjustments did not solve the problem.

Simple instruments that require manual measurements have

been tried by the authors with varying degrees of success,

because several possible sources of error may be introduced.

Therefore, the authors generally avoid instruments that require

manual measurements except where necessary (such as closure

measurements).

Instruments are available from several manufacturers. The

authors have used vibrating-wire sensors, and linear and rotary

potentiometers. Variations on these instruments are available

from Geokon, Inc., Lebanon, NH; Roctest, Inc., Plattsburgh,

NY; RST Instruments, Vancouver, British Columbia, Canada;

and Slope Indicator Co., Bothell, WA. SRL researchers have

used string (rotary) potentiometers made by Celesco Transducer

Products, Inc., Canoga Park, CA, with hardware manufactured

in-house to monitor sag at different depths in a coal mine roof

[Signer et al. 1993].

INSTRUMENTED BOLTS

An instrumented bolt [Serbousek and Signer 1987; Johnston

and Cox 1993] (figure 5) is an instrument that has been used

increasingly in recent years by SRL researchers [Signer 1990,

1994; Larson et al. 1995; Larson and Maleki 1996]. It consists

of strain gauges bonded to milled surfaces on opposite sides of

a bolt. In addition to measuring strain, the gauges may be

calibrated in the elastic range to calculate load directly as long

as load has not progressed past the yield point of the bolt steel.

If gauges are monitored over time, a row of these instrumented

bolts reveals useful information about roof deformation and

roof conditions. The information may aid in determining

potential roof hazards and justifying changes in roof support to

optimize or increase safety. The instruments are commercially

available from Jenmarr Corp., Pittsburgh, PA, and Rock

Mechanics Technology, Ltd., Staffordshire, U.K., although the

authors have only used bolts instrumented by SRL technicians.

Because of the low level of the signal, the potential of the

instrument serving as a ground, and differences in ground

potential from point to point in an underground roof, external

noise may be introduced to the gauge signal. The user should

follow suggestions found in the "Data Acquisition System

Basics" section to minimize the likelihood of this problem.

If long cables link the gauges to a data acquisition system, a

three-wire, quarter-bridge circuit should be used.

INSTRUMENTED CABLE BOLTS

A new development in cable bolt sensor technology is replacement of the bolt's original king wire with an instrumented

king wire [Martin et al. 2000] (figure 6). Similar to an instrumented bolt, an instrumented cable bolt is a support device

as well as a rock mass strain-measuring device in underground

mines. Characteristics of the rock mass in an underground

mining environment can be determined by these devices, which

provide a means for mining engineers and mine inspectors to

predict roof falls.

The gauges are mounted on a narrow strip of steel at

intervals determined by expected roof conditions, the instrumented metal strip is placed in a steel mold, and the mold is

injected with epoxy. This results in a new king wire. The

remaining six cables are then rewound around the new king

wire, creating a seven-stranded instrumented cable. Maximum

length of the cables is 6 m (20 ft) because of the physical

constraints of the wires connecting the gauges. The maximum

number of gauges is 10, or 5 on each side.
261

 Figure 7.—Hollow inclusion stress cell. A, Strain

gauge shell with piston, dowel, and cable; B, closeup of

shell and piston.

The gauges can either be set in pairs in the same plane

to detect shear or at 90E angles to detect lateral loading.

The cables are inserted in a mine roof using cement or resin

grout. They are capable of holding an ultimate load of 214 kN

(48,000 lb). Data readings are reliable to bolt loads of 178 kN

(40,000 lb).

The cables are fitted with a special hex head and inserted

into the rock with either a jackleg drill or a conventional roof

bolting machine. The cable bolts are spun into the drill hole,

mixing the grout for adhesion. The special head also has a

connector that allows for data acquisition after installation.

The same cautions and limitations that apply to instrumented

bolts apply to instrumented cable bolts.

HOLLOW INCLUSION CELL

Worotnicki and Walton [1976] developed the hollow

inclusion stress cell (HICell) (figure 7) at the Commonwealth

Scientific and Industrial Research Organization (CSIRO) in

Australia. These cells are available from Mindata, Seaford,

Victoria, Australia, and from Reliable Geo, LLC, Yakima, WA,

in the United States. The HICell is actually a group of strain

gauges and strain gauge rosettes embedded in a thin epoxy

shell. It is bonded to the wall of an EX-size borehole.

Installers drill to the approximate installation depth with a 152-

mm (6-in) barrel. An EX hole is extended for approximately 61

cm (24 in). Glue is displaced by a piston and forced into the

annulus between the shell and the borehole wall. When stress

is relieved through overcoring [Hooker and Bickel 1974], the

stress state may be inferred from strain measurements collected

from one installation.

Isotropic elastic properties are determined using results of a

biaxial test on a recovered rock core and instrument. For

isotropic rock, Larson [1992] developed a data reduction code,

STRESsOUT. For transversely isotropic rock, five independent

elastic constants are needed, and Amadei [1983] has developed

a data reduction solution. However, laboratory tests on cores

with orientations along the axes of anisotropy are required to

determine all elastic constants. These constants can be

estimated, but such estimates decrease the quality of stress state

determination.

Tesarik and McKibbin [1989] installed and overcored four

HICells at the Magmont Mine, a lead mine in Missouri. The

stresses they determined from the overcoring measurements

were subsequently used as initial stresses in a two-dimensional,

finite-element model employed to analyze stress redistribution

during pillar recovery.

Johnson et al. [1993] overcored four HICells and a borehole

deformation gauge in three boreholes at one site and two

boreholes at another site in the Homestake Mine, Lead, SD.

Because results from overcoring varied, they discussed causes,

such as variability of geology and material properties,

anisotropy, and use of statistics with the least-squares approach.

In an underground platinum and palladium mine, Johnson

et al. [1999] used HICells to measure in situ stress and

subsequent stress change ahead of the face during advance of

a footwall drift. Stress components calculated by multiple

linear regression fit the strain measurements with little error.

The correlation coefficients for each determination were in the

range of 0.90 to 1.00.

Usually it is difficult to obtain good results in coal-measure

rock because most sedimentary cores fracture easily during

drilling. However, if good cores can be obtained, overcoring

the HICell will usually result in good-quality measurements.

Conducting a dry run before HICell installation is very

important. This is accomplished by using nails instead of lead

pins in the shear pinholes. This practice allows an installer to

learn to sense the right technique for inserting an instrument in

an EX borehole.

A problem was recently encountered while installing a

HICell. Three seal rings are positioned close together on the

base of the instrument and on the piston. Recently, it was found

that the grooves for those seals were positioned closer together

than had been done in the past. In a 38.1-mm (1.500-in) hole,

these seals did not have room to bend and thus were forced over

the top of each other. This made it impossible to push the

piston into the EX hole. A solution discovered by the SRL

installation team and verified by Mindata [Walton 1999] was to

remove the middle seal on both the base of the instrument and

the piston. We have completed two successful installations

with this modification. Also, Mindata now recommends not

cutting V-shaped grooves in the seals, as is encouraged in the

installation manual.
262

Figure 8.—Borehole deformation gauge.

 Figure 9.—Schematic of three-component borehole deformation gauge. A, Main body showing location of taperedmounted cantilevers and pistons; B, closeup of a cantilever showing location of strain gauges (after Bickel [1993]).

BOREHOLE DEFORMATION GAUGE

The three-component borehole deformation gauge (figure 8)

was designed by the former U.S. Bureau of Mines (USBM)

[Merrill 1967]. Pistons that contact the borehole wall press

against cantilevers mounted inside the body of the instrument

(figure 9). Strain gauges bonded near the base of the cantilevers

(two on the top surface and two on the lower surface) are

connected as a full Wheatstone bridge and calibrated to reflect

changes in the diameter of the borehole. Like the HICell, this

gauge is installed in an EX hole and overcored [Hooker and

Bickel 1974; Bickel 1993]. During overcoring, borehole diameter

changes are measured in three different orientations 

spaced at 60E intervals. Pistons that contact the borehole wall

displace a cantilever. Strain gauges bonded near the base of 

the cantilever form a full Wheatstone bridge circuit. Changes

in signal are calibrated to calculate changes in diameter. After

overcoring, a biaxial test is performed to measure the Young's

modulus of the overcore.

The borehole deformation gauge is easy to use and install.

There is no curing time as there is for the HICell. However,

much drilling is required to determine all independent

components of the stress tensor because the instrument must be

overcored three different times, each time in a different noncoplanar borehole. Stress components are fit to the measurements using the least-squares method, as demonstrated by

Panek [1966]. This calculation may be performed with the

STRESsOUT code [Larson 1992].

NIOSH researchers have recently used this instrument for

measuring horizontal stress in a coal mine roof [McKibbin

2000]. In this case, one borehole orientation was used. The

results from multiple overcores were consistent. The method

and data reduction are described by Bickel [1993], but in his

analysis, he assumed that the borehole axial strain was zero

because it was not measured. That assumption must be

evaluated. Equations used in data reduction are discussed by

Obert and Duvall [1967].
263

Figure 10.—Biaxial stressmeter (from Seymour et al. [1999]).

Girard et al. [1997] overcored a borehole deformation gauge

at the 7400 level of the Homestake Mine. The stress state

determined from these measurements compared well with

stresses calculated from equations developed by Pariseau

[1986], which represented principal stresses as a function of

depth. Measurements by Johnson et al. [1993] in conjunction

with HICell measurements were mentioned in the previous

section.

The borehole deformation gauge is available from Geokon.

BIAXIAL STRESSMETER

The biaxial stressmeter is a rugged and reliable stress monitoring instrument manufactured by Geokon. The stressmeter

is typically grouted in a BX-size (60-mm (2.36-in)) drill hole to

measure compressive stress changes in a plane perpendicular to

the instrument's longitudinal axis. As the walls of the drill hole

deform in response to stress increases induced by mining,

deformation in the host rock is transferred through a thin,

grouted annulus to the stressmeter's body (a thick-walled steel

cylinder approximately 34 cm (13 in) long). Radial deformation of the steel cylinder is measured by one or two sets

of three vibrating-wire gauges oriented 60E from each other and

closely spaced along the center of the instrument's longitudinal

axis (figure 10). From these radial deformation measurements,

the magnitude and orientation of changes in major and minor

secondary principal stresses can be calculated using equations

developed by Savin [1961]. The biaxial stressmeter can also be

equipped with two longitudinal vibrating-wire gauges that

measure overall change in the length of the instrument. To

account for the effects of stress changes outside of the biaxial

measurement plane, the longitudinal measurements are used to

calculate axial strain and correct the computed secondary

principal stress changes for the Poisson component of stress

changes acting in the direction of the drill hole. In addition, the

stressmeter can also be equipped with two vibrating-wire

temperature sensors that measure the temperature at the

instrument's location. Because the vibrating-wire gauges have

a slight temperature sensitivity, the manufacturer provides

temperature calibration factors for each of the radial and

longitudinal gauges. Data from the temperature sensors are

used in conjunction with these calibration factors to correct the

computed radial and longitudinal strains and compensate for the

influence of temperature fluctuations near the instrument.

The manufacturer currently supplies various models of the

biaxial stressmeter with either three or six radial gauges and

with or without the longitudinal gauges and temperature

sensors. Because consistent readings from all three of the radial

gauge orientations (i.e., 0E, 60E, and 120E) are required to

calculate the magnitude and orientation of the secondary
264

Figure 11.—Borehole pressure cell.

principal stress changes, the authors highly recommend using

a duplicate set of radial gauges for data redundancy and added

reliability. Although the amount of longitudinal and temperature corrections have generally been small, the authors also

prefer to order the biaxial stressmeter with two longitudinal

gauges and two temperature sensors so that these corrections

can be quantified accurately and the stress changes adjusted if

necessary. For situations where maximum accuracy is desired

or the temperature changes are expected to be large, the instrument should be ordered with the full complement of radial,

longitudinal, and temperature gauges.

Before installing the biaxial stressmeter, the authors typically

perform a point-load test with the instrument to confirm internal

wiring configurations, verify the operation of the radial gauges,

and check the configuration and programming of data acquisition equipment. Usually the electronic readings collected

from the biaxial stressmeter are analyzed using spreadsheet

software because of the complexity and number of datareduction equations. As a general rule, the response of the

vibrating-wire transducers should be graphed and analyzed

prior to converting the electronic readings to engineering units

and calculating the magnitude and direction of the stress

change. Graphing the electronic readings helps identify

anomalous data, diagnose the performance of the datalogger

components, and interpret the response of individual gauges

that are not functioning as expected. Inconsistent data can be

difficult to detect and interpret from stress change calculations

alone.

To measure stress changes in a vertical plane, the biaxial

stressmeter is typically installed in a drill hole inclined slightly

below horizontal to ensure that the grout flows to the bottom of

the hole and completely fills the annulus around the instrument.

The installation hole is usually drilled with a diamond core drill

for about 1.5 m (5 ft) past the target depth for the instrument.

After initially filling the hole with about 3 m (10 ft) of grout,

the biaxial stressmeter is inserted into the drill hole, pushed

through the grout to the target depth, and carefully rotated into

position (i.e., radial gauge 1 oriented vertically) using selfaligning setting rods or a leveling device. To lock the instrument securely at this desired position in the drill hole, two

opposing snap-ring anchors are activated from the collar of the

hole by pulling a cable attached to an anchor-release pin.

The remainder of the hole is then filled with grout to help

ensure that the biaxial stressmeter is completely covered with

grout. Grouting the entire hole also alleviates problems with

stress concentrations or creep in the drill hole near the

instrument's location, which can introduce errors in the stress

change measurements. The biaxial stressmeter is generally

installed using a relatively thick mixture of Five Star Special

Grout 400, an expansive cement-based grout manufactured by

Five Star Products, Inc., of Fairfield, CT. A thick grout mix

helps deter grout migration through fractures or discontinuities

near the drill hole and provides sufficient viscosity to displace

extraneous water from the hole.

Using these procedures, Seymour et al. [1999] have installed

biaxial stressmeters in a variety of underground mines and host

rock formations to monitor and evaluate mining-induced stress

changes in mine pillars and abutments. By measuring both the

magnitude and direction of the secondary principal stress

changes, the biaxial stressmeter provides valuable insights into

the behavior of the host rock and the redistribution of mininginduced stresses. Data from these instruments have been used to

determine the transfer of mining-induced loads to cemented backfill pillars [Tesarik et al. 1991], evaluate the ground support

capability of frozen gravel pillars [Seymour et al. 1996], and

monitor stress redistribution ahead of a longwall [Seymour et al.

1998]. The instrument's robust design and vibrating-wire transducers are particularly well suited for monitoring long-term

changes in stress. Because the biaxial stressmeter is securely

grouted in the host rock, the instrument does not seem to be adversely affected by vibrations caused by blasting or mining

equipment.

To measure stress changes in a horizontal plane, the biaxial

stressmeter is installed in a vertical drill hole. At a proposed

nuclear waste repository in Sweden, several biaxial stressmeters

have been installed in boreholes drilled in the bottom of a test

drift to measure both thermal and mining-induced stress

changes around cylindrical excavations that may eventually

house canisters containing radioactive waste material [Röshoff

1999]. In addition, Geokon is currently developing a borehole

packer for installing the biaxial stressmeter in overhead drill

holes. SRL researchers recently used a prototype version of

this borehole packer to install biaxial stressmeters in the roof of

a coal mine.

BOREHOLE PRESSURE CELLS

Borehole pressure cells are used to monitor stress change in

one direction (e.g., vertical stress change profiles at several

points through a pillar). They consist of rectangular flatjacks

embedded in a cylinder of grout (figure 11). The cylinder is

placed in a 60-mm (2.375-in) diameter borehole and set to a

pressure about 10% more than the estimated absolute stress in

the host rock. The viscoelastic-viscoplastic nature of the coal

or rock allows a pressure equilibrium to be reached. Thereafter,
265

Figure 12.—Earth pressure cell.

any changes in stress of the host rock are reflected directly in

equal changes in fluid pressure in the flatjack.

These gauges may be monitored with hydraulic gauges.

Typically, we have used both hydraulic gauges and pressure

transducers and monitored them with a datalogger.

Three cells may be installed in the same borehole. After a

cell is placed and oriented in the borehole, the cell is seated by

pumping hydraulic fluid into the flatjack. The cell is then

pressurized to a predetermined level before the valve is turned

off. Resetting the pressure is sometimes necessary after 24 hr.

The hydraulic failure rate appears to be about 10% to 20%

when the cells are manufactured by experienced hands. Goodquality cells can only be achieved if made by a careful,

experienced manufacturer. The authors used cells manufactured only by employees of the Denver Research Center of

the former USBM. However, Geokon manufactures a flatjack

intended to be grouted into a borehole.

EARTH PRESSURE CELLS

Hydraulic earth pressure cells (figure 12) are commonly

used to measure stress changes in unconsolidated or cemented

materials. This instrument consists of two rectangular or

circular stainless steel plates welded together around their

edges. A thin cavity is left between the two plates and filled

with a fluid such as antifreeze or hydraulic oil. Pressure applied

to the cell induces an equal pressure on the internal fluid that,

in turn, is sensed by a transducer connected to the cavity

between the two welded plates with high-pressure stainless steel

tubing [Dunnicliff 1988].

Earth pressure cells are generally used to identify loading

trends and are not relied upon for precise measurements.

Because the stiffness of the instrument and the medium in

which it is placed most likely will not be equal, the earth

pressure cell may not sense the same stress change as the

surrounding medium. Furthermore, the modulus of the in-place

material may vary considerably from one location to another as

a result of unequal compaction or heterogenous components,

such as coarse or fine aggregate. This modulus variability may

cause discrepancies or scatter in the stress change measurements. Finally, it is difficult to calibrate the instrument accurately in the laboratory in a medium other than fluid.

The authors generally use earth pressure cells with vibratingwire transducers to measure stress changes in cemented

backfill. To avoid point loading or damaging these instruments

when they are installed in cemented rockfill, the earth pressure

cells are precast in wood forms using a cemented backfill mix

with minus 6.35-mm (0.25-in) aggregate. The forms are

constructed large enough to form a protective layer several

inches thick on all sides of the instruments. As soon as the cast

instruments can be handled without breaking the protective

layer, the forms are removed, and the instruments are placed at

their desired orientation in the backfill stope. Casting the

instruments helps maintain alignment during installation and

protects the cells when wet backfill is placed over them. To

ensure that the desired orientation of the earth pressure cells is

maintained during the bulk filling of an underground opening,

an additional protective layer of cemented rockfill

approximately 0.6 m (2 ft) thick is placed over the instruments

and allowed to cure for at least 24 hr before heavy equipment

is driven over the in-struments. Earth pressure cells installed

using these procedures have supplied important information

regarding the mining-induced loads that were transferred to

cemented backfill stopes as adjacent ore panels or pillars were

mined [Tesarik et al. 1993, 1995]. They have also been used to

evaluate the long-term ground support provided by cemented

backfill [Tesarik et al. 1999].

In mines that use hydraulic fill, earth pressure cells are

anchored to wire mesh using cable ties or bailing wire. If wire

mesh or some other means of anchoring the instrument is not

available at the desired location in the stope, the instruments

can be suspended with wire in a wooden frame without a top or

bottom and placed directly in the fill by hand. However, care

should be taken to maintain the desired orientation of the

instrument. This technique was successfully used at an

underground coal mine; the earth pressure cells were installed

in in-panel entries as they were backfilled with an air-entrained

mixture of fly ash and cement [Seymour et al. 1998].

Earth pressure cells are made by several instrument manufacturers, including Geokon, Roctest, RST Instruments, and

Slope Indicator Co.

EMBEDMENT STRAIN GAUGES

The authors generally use embedment strain gauges (figure 13) with vibrating-wire transducers to measure relative

displacement in cemented backfill. Strain is calculated by

dividing the relative displacement between the instrument's

two end flanges by the length of the instrument, typically 15 to

25 cm (6 to 10 in). Typically these instruments are installed
266

Figure 13.—Embedment strain gauge.

near the location of an earth pressure cell to provide an estimate

of the backfill's in situ modulus (i.e., stress change divided by

strain change).

For cemented backfill with a modulus ranging between 690

and 1,380 MPa (100,000 and 200,000 psi), a 25.4-cm (10-in)

long embedment strain gauge with a 6.35-mm (0.25-in)

maximum displacement range can be used to determine strain

changes up to a stress change level of between 17 to 34 MPa

(2,500 to 5,000 psi). Problems have been encountered using

embedment strain gauges with smaller displacement ranges

because the maximum displacement range was exceeded. On

the other hand, erratic results were obtained from 1-m (39-in)

long instruments with 12.7-mm (0.5-in) displacement ranges

that were placed vertically in cemented fly ash. The reason for

these variations is not fully understood, but the embedment

strain gauges may have buckled as mining-induced loads were

transferred to the backfilled entries because of the longer length

of the instruments.

Installation procedures for embedment strain gauges are the

same as for earth pressure cells, except waxed cardboard

cylinders are typically used to precast the instruments instead

of wooden forms. Except for the above-mentioned problems,

these instruments have generally provided good results and

have typically given a more precise indication of backfill

loading changes than the earth pressure cells.

Embedment strain gauges are made by Geokon, Roctest, and

Slope Indicator Co.

VERTICAL BACKFILL EXTENSOMETERS

To measure relative displacement over large vertical

distances in backfill, vertical extensometers can be constructed

as shown in figure 14. For applications where boreholes can be

drilled in placed backfill, sections of 15.9-mm (0.625-in)

diameter steel rod coupled together serve as the extensometer

measurement rod, and steel rebar welded to a small length of

rod is used as the downhole anchor. A mixture of cement and

water is poured down the hole to set the anchor, followed by

dry sand to prevent pieces of backfill from spalling and

wedging between the steel rod and the side of the drillhole. The

head assembly, which consists of a vibrating-wire displacement

transducer, bearing plate, and a short length of 38-mm (1.5-in)

steel pipe welded to the plate, is placed over the steel

measurement rod, and the transducer is screwed into the rod.

The transducer is set at the desired position using an aluminum

extension rod and is held in place with a compression fitting.

Finally, the excess portion of the aluminum rod extending

above the compression fitting is cut off, and a protective head

is bolted onto the bearing plate. These instruments have

successfully monitored compressive strains in backfill caused

by overburden stress redistribution, as well as tensional strains

caused by undercutting backfill [Tesarik et al. 1993].

When drilling in cured backfill is not feasible, extensometers

can be assembled as successive backfill lifts are placed. In this

situation, the bottom anchor is constructed of a 45.7- by

45.7-cm (18- by 18-in) steel baseplate with a 51-mm (2-in) steel

coupler welded to the top of the plate. The first section of a

15.9-mm (0.625-in) diameter steel measurement rod is inserted

through a hole in the baseplate and secured with a nut. The first

section of 51-mm (2-in) pipe is threaded into the steel coupler

to protect the measurement rod as backfill is placed around the

instrument. The outside surface of this pipe is greased during

construction to help reduce friction between the outer surface

of the pipe and the cured backfill. As successive lifts of backfill

are placed, sections of rod and pipe are coupled together until

the desired instrument height is achieved. The 38-mm (1.5-in)

steel pipe on the head assembly is then inserted inside the

51-mm (2-in) protective pipe, leaving several inches of backfill

between the bearing plate and the top of the 51-mm (2-in) pipe.

This gap allows the instrument to retract as the backfill

compresses under load. The transducer is set and secured at the

desired position as described above, and the protective cover is

bolted onto the head assembly. Following these procedures,

backfill extensometers ranging from 4.3 to 16.5 m (14 to 54 ft)

long have been successfully installed in cemented fill stopes

and used to determine the vertical strain induced within the

backfill by mining [Tesarik et al. 1991, 1995].

HORIZONTAL BACKFILL EXTENSOMETERS

The authors have attempted to measure horizontal strain in

cemented backfill stopes (horizontal dilation in response to

vertical loading), with only marginal results. At one mine site,

two 3-m (10-ft) long commercially available borehole joint

meters were hung end-to-end from steel cables that spanned the

width of a 7.3-m (24-ft) wide stope [Tesarik et al. 1991]. Some

of these instruments were damaged during installation when

backfill dumped from the bench above surged forward and

distorted the flexible instrument into a slight arc. Data recorded

by the jointmeters that still functioned after the stope was

completely filled were difficult to interpret because, unlike most

of the other rock mechanics instruments installed at this site,
267

Figure 14.—Vertical backfill extensometer.

changes in instrument readings were not always related to a

specific mining event. Compressive strain, which may be

attributed to backfill shrinkage, was recorded by some of these

instruments after the stopes were first filled.

To help prevent damage from surging or falling backfill,

horizontal extensometers were fabricated at SRL with steel,

rather than plastic, protective pipe. Steel-bearing plates at each

end of the extensometer were also enlarged to increase the

contact area between the instrument and the backfill. These

instruments were placed directly on the floor of partially

backfilled stopes and covered with wet backfill using a frontend loader. This practice protected the instruments from the

weight of mining equipment and the large backfill loads that

were subsequently placed over the instruments. These more

robust extensometers recorded horizontal compressive trends

from 73 to 145 days after installation, followed by relatively

small tensional strain changes that were difficult to interpret

[Tesarik et al. 1993].
268

CONCLUSION

Design and implementation of a successful instrumentation

plan requires careful planning and consideration of site-specific

factors, such as geology, identification of critical mechanisms

and parameters, clearly defined objectives of measurements,

estimates of the range of these measurements, types of

transducers, ease of installation, how the sensors will be

monitored and monitoring interval(s).

An overview of sensor technologies has been presented.

Precision and the ability to sense sudden changes is a distinct

advantage of Wheatstone bridge technology, but zero drift often

makes it unsuitable for long-term monitoring. Vibrating-wire

technology has been proven to provide reliable long-term

measurements. Potentiometers have been used successfully to

measure displacement.

The authors have summarized the use of several instruments

and offered tips to help the user obtain better measurements.

Also, the experiences of coworkers are cited. It is hoped that

the information conveyed here will enable mine operators to

make measurements on their own or improve on past

measurement programs and that measurements obtained might

be useful in making decisions that improve the safety of miners.

ACKNOWLEDGMENTS

The authors express their gratitude to John McRae of

Geokon, Inc., Lebanon, NH, and Jeff Goalen of Campbell

Scientific, Inc., Logan, UT, for reviewing this paper and

offering corrections and suggestions for improvement.

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271

EVALUATION OF GROUND SUPPORT AT A TRONA MINE

 USING INSTRUMENTED CABLE AND REBAR BOLTS

By Lewis A. Martin,1 Catherine A. Dreesbach,2 and Rimas Pakalnis, Ph.D.3

ABSTRACT

Instrumented cable bolts developed at the Spokane Research Laboratory of the National Institute for

Occupational Safety and Health were used in conjunction with existing ground control systems to monitor rock

mass loads at Tg Soda Ash's trona mine, Granger, WY. Axial and shear loads were determined to levels of

strain gauge accuracy of ±5 N or ±5 microstrain. These gauges were embedded in a remanufactured king wire

that replaced the conventional king wire. Cable bolt performance, quality of grout, and installation techniques

were also assessed. By using instrumented cables, a mine operator can determine axial load along the cable

at predefined gauge locations. These data provide necessary input for an operator to design a safer working

environment for miners.

1

Mechanical engineer, Spokane Research Laboratory, National Institute for Occupational Safety and Health, Spokane, WA.

2

Mine engineer, formerly with FMC Granger, Granger, WY.

3

Associate professor, Department of Mining and Mineral Process Engineering, University of British Columbia, Vancouver, British Columbia, Canada.
272

 Figure 1.—Calibrated curve for instrumented king wire cable

bolt.

 Figure 2.—Instrumented cable bolt assembly with and

without pretension head.

BACKGROUND

Until recently, external measurement devices such as the

"tensmeg" or resistance-wire cable strain gauges [Hyett et al.

1997; Windsor 1987] were used to determine load on cable

bolts. Previously, researchers thought that an internal gauge

would be too expensive to use with cables as an instrument

within mines. Another issue has been that external gauges on

cables may fail because of water seepage through the grout

before the cable begins to take on load [Goris et al. 1993]. In

1996, the "stretch measurement to assess reinforcement

tension" (S.M.A.R.T) cable was developed and patented by

researchers at Queen's University in Canada [Hyett et al. 1997].

This system replaces the king wire of a cable bolt with

multipoint extensometers enclosed in a tube to measure cable

elongation, 

thereby enabling deformation and load to be determined

[Bawden et al. 1998]. The usefulness of a ground control

instrument having gauges mounted in the center of the cable or

on the king wire was shown by Signer [1990] in research on the

use of resin-grouted rebar bolts in coal mines.

In 1998, researchers at the Spokane Research Laboratory

(SRL) of the National Institute for Occupational Safety and

Health (NIOSH) developed an instrumented king wire cable

bolt (patent pending). This instrument is manufactured by

molding the king wire from epoxy around a flat piece of steel

containing strain gauges. Instrumented cable bolts will aid in

evaluating geologic conditions, grout quality, effective

anchorage length, and cable load.

CONCEPT

The instrumented king wire cable bolt has 12.5-mm-long

strain gauges embedded in the king wire at positions defined by

the mine operator. Presently, 10 gauges can be installed along

a resin-grouted bolt up to 5 m long or a cement-grouted bolt up

to 7 m long. These constraints are largely due to current

manufacturing processes and not to limits on the system. The

gauges can either be set in pairs to detect shear load in the same

plane, or the flat stock can be set at 90E angles to detect loads

in three directions. The cables are capable of sustaining an

ultimate load of 250 kN and can produce reliable readings to

180 kN, as shown by the calibration curve in figure 1.

To create an instrumented king wire cable bolt, strain gauges

are first installed on both sides of a 2.4- by 0.5-mm metal strip

at various user-specified positions on the bolt (figure 2). The

individual strain gauges are connected to an instrument plug by

30-AWG wire. This apparatus is then placed in a steel mold

and injected the total specified length of the bolt with a two-part

epoxy. This new king wire measures 8.3 mm in diameter.

The original cable is uncoiled strand by strand, the original

center wire is replaced by the new epoxy-filled king wire, and

the outer six strands are rewound around it, resulting in an

instrumented cable bolt. The gauge wires protruding from the

center are then inserted into a 12-pin connector. This connector

is recessed into a 4.4-cm hex head attached to the cable by

means of a barrel-and-wedge assembly (figure 2). The cables

are fitted with special heads so they can be inserted by either a

jackleg drill or a roof bolter. An instrumentation plug is used

to monitor the gauges. A data acquisition system is required to

read the loads.

The system has been tested on a Vishay strain indicator box,

a Campbell Scientific system, and an Omni Data system.
273

 Figure 3.—Calculated load and strain on grouted cable bolt with instrumented king

wire.

MATERIAL DESCRIPTION

Standard cable specifications were the American Society of

Testing Materials (ASTM) standard A416, 0.6-in, grade 270 K,

low-relaxation, seven-strand cable having a minimum breaking

load of 258 kN with 3.5% elongation. Upon removing the

original king wire and replacing it with the manufactured one,

ultimate load was reduced to 215 kN. Strain readings were

reliable to 180 kN with 3.5% strain on the cable at failure.

Calibration tests at SRL provided correlation of microstrain to

load to enter into the data collection system (figure 1). This

technique was used successfully throughout the elastic loadstrain range on instrumented rebar bolts by Signer et al. [1997].

CABLE DESIGN

Laboratory tests were conducted on instrumented king wire

cable bolts embedded with resin grout in concrete blocks to

obtain calibration curves for converting microstrain to load in

kilonewtons. Pull tests were also conducted to determine the

load-carrying characteristics of the bolts. The strain gauges

were mounted 0.15, 0.35, 0.61, 0.91 and 1.22 m from the head

of the bolt (figure 3). The cables were loaded to 180 kN using

hydraulic pull test equipment. A strain-to-load curve is shown

in figure 3 and is typical of what Goris et al. [1994] observed in

pull tests with conventional cable bolts.

These tests aid in explaining how load on cable bolts is

transferred from the cable to the grout and then into the rock

mass. A "shelling" effect is shown whereby the grout-cable

interface breaks down and the gauge becomes debonded as load

is increased and is taken fully by the cable. Figure 3 shows

results of the pull tests as a load profile along the cable length.

The curve shows that a 128-kN (~12.8-tonne) pull load on the

collar of the cable resulted in measured strain loads of 123 kN

along gauges 1 and 6 at a distance of 0.15 m from the collar.

Researchers interpreted this measurement, a difference of 5 kN,

as the amount of load the cable was not sensing 0.15 m from the

collar. At 1.22 m from the collar, the cable did not sense any

load, indicating that 1.22 m was the critical bond length. Other

researchers (e.g., Goris et al. [1994]) have reported critical bond

length to be approximately 1 m for standard cable. This load

capability enables an operator to measure failure in real time as

debonding of the grout-cable interface is initiated. In addition,

the difference between cable load and applied load indicates

grout quality, which can be a better field test of the quality of

cable bolt installations than field pull tests.
274

 Figure 4.—Cross section of resin-grouted rebar

bolt machined for installation of gauges.

 Figure 5.—Typical load strain profile for instrumented

rebar bolt.

REBAR DESIGN

The resin-grouted rebar roof bolts used in these tests were

standard, grade 60, No. 6 rebar that had been milled with a

6.35-mm-wide by 3.18-mm-deep slot along each side (figure 4).

Axial load tests showed that the average yield load of the bolts

was 96 kN and the average ultimate load was 158 kN. Before

milling the slot, yield load had been 107 kN and ultimate load

had been 176 kN. Thus, slotting caused a 10% reduction in

strength [Signer 1990].

Strain gauges were installed into the slots as shown in figure 4. The strain gauges were then calibrated in a uniaxial test

machine to correlate voltage change to load change. A typical

load-strain curve from a tested bolt is shown in figure 5. When

the data from the instrumented rebar bolts were reduced, the

correlation coefficients from the axial calibrations were used to

convert voltage changes to load changes. This process was

accurate to ±22 N. When the load levels exceeded the yield

point of the steel, voltage readings were converted to strain

readings to estimate the degree of plasticity (figure 5).

CASE STUDY: TG SODA ASH, FMC GRAINGER

MINE GEOLOGY

The mine is located in the Green River Basin of southwestern Wyoming. Trona deposits in the basin were formed

during the Eocene period within the Wilkins Peak Member of

the Green River Formation. The repeated evaporation and

recharge of Lake Gosiute, a playa lake occupying the basin

during Eocene time, created evaporite layers of trona

intermixed with marlstone and alternating layers of oil shale

[Leigh 1998].

The areal extent of the basin is marked by the Uinta

Mountains in the south, the Rock Springs uplift in the east, the

Wind River and Gros Ventre ranges in the north, and the

Wyoming Range in the west. There are 42 known trona beds,

of which 25 range from 240 to 670 m thick. These beds are

numbered 1 through 25, according to the order in which they

were deposited (bed 25 being closest to the surface). Beds 1

through 18 consist mainly of fine-grained trona; beds 19

through 25 are composed of a coarse crystalline trona. Mining

at Tg Soda Ash takes place in bed 20 approximately 425 m

below the surface. The area in which the field study was

conducted is overlain by an additional 60 m of overburden.

MINING METHOD

Trona is mined with a drum-and-borer-type continuous

mining machine. These machines, traditionally used for coal

mining, are custom built to cut the harder trona ore. Tg Soda

Ash uses only drum-type continuous miners. Cutting heads are

typically 5.3 m wide, but the panel in which the field study was

conducted was mined with a 5.6-m-wide cutting head.

The Tg Soda Ash mine uses a modified room-and-pillar

mining method. In the field study area, the panel was

developed as a four-entry system. An additional two entries

were later driven parallel to the existing workings, with an

eight-room retreat mined in a chevron configuration 78 m wide

by 103 m deep with cross cuts on 12.2-m centers. The retreat

consisted of pulling chevrons, a method in which pillars are cut

on 12.2-m centers 60E from the direction of advance. This

method maximizes the rate of extraction by minimizing ground

control problems.

The field study site was chosen for its rapidly changing

ground conditions and its proximity to mining activity in this

panel. It is believed that the wider cutting head, extra overburden, and weaker geologic material of the roof and floor in
275

Figure 6.—Instrument locations in mine. Figure 7.—Bolt and sagmeter locations.

the panel contributed to rapidly deteriorating ground conditions

in the outside entries of the system, making the area ideal for a

field study. During the course of the study, the face was

approximately 215 m from the test area and retreated to within

45 m of the test area.

INSTRUMENTED BOLT INSTALLATION

For simplification in the rest of this paper, the term "cable

bolt" means a bolt in which the conventional king wire has been

replaced with the new remanufactured king wire.

Cable bolts and rebar bolts were installed in the roof of the

mine to measure roof loads in the intersection of 350S and

12371E (figures 6 and 7). This location was chosen because of

drill access ahead of the mining cycle and because loads would

be induced in the bolts as ore was extracted toward the

intersection. Mine personnel were interested in determining the

potential for instability in the back and the extent of any

unstable blocks.

Ground support was installed after development and during

rehabilitation. Strain gauges were positioned along rebar and

cable bolts at depths that would ensure that loads on the resingrouted support would be measured. In addition, a set of

gauges was mounted at an ungrouted section to measure deadweight loads in the rock mass. Figure 8 shows the location of

the strain gauges with respect to the collar. The longest cable

bolts were installed at the center of the intersection because

these bolts would receive the greatest loads, while shorter cable

bolts were installed north of the intersection closer to the pillar

front. Resin-grouted rebar bolts were placed in the south part

of the intersection to measure loads where the density of rebar

bolts was highest. Sagmeter stations were set at three locations

in the intersection (figure 7), and holes were drilled to depths of

2.4, 3.2 and 3.8 m. These sagmeter depths were chosen to

correspond with the bolt gauge profiles (figure 8), allowing

correlations to be made between deformation recorded by the

strain gauges to deformation recorded by the sagmeter stations.

This is important because slip expected to occur between the

cable-grout-rock interface can be compared to actual rock

deformation.

RESULTS OF MEASUREMENTS ON

INSTRUMENTED BOLTS

Loads on the cable bolts were measured from April 29

through July 11, 1999. Mining activities influencing the

intersection progressed from northeast of the site to the south

(figure 6). On June 9, development mining began within one

pillar width (8 m) of the intersection. Figure 9 shows all load

profiles measured by the 4.27-m-long bolts (bolts 1, 2, 4, and 5)

and the 3.66-m-long bolts (bolts 6, 8, 9, and 11). Approximately the last 1.5 m of the bolt was bonded with resin

grout. Figure 10 shows load profiles for the 2.44-m-long, fully

grouted rebar bolts (bolts 201-204).
276

 Figure 8.—Positions of strain gauges on bolts (note two

at each position).
277

 Figure 9.—Load profile versus time on 4.27-m-long bolts (1, 2, 4, and 5) and 3.66-m-long bolts (6, 8, 9, and 11).
278

Figure 10.—Load profile versus time on a 2.4-m-long, fully grouted rebar bolt.

Following is the authors' interpretation of loads on 12 bolts. By

recording actual point loads in the bolts, it will be possible to

understand more readily the likelihood that instabilities may be

forming or that there may be problems with the quality of the

installation.

$ Gauges 2 and 7 on cable bolt 1 indicated loads up to the

yield limit of the steel. These loads were higher than the axial

dead-weight loads on the ungrouted section of the bolt and

loads on the grouted gauges, indicating that some delamination

was occurring close to the gauge. The higher readings were

most likely a result of the fact that concentrated load is

measured over the gauge length by a strain gauge, whereas an

ungrouted gauge measures all delamination cracks over the

length of the ungrouted section, which was 2.8 m. The result is

an average load much lower than the load on a bonded cable

over a given distance.

$ Bolt 2 was inspected by mine personnel, who determined

that it had broken. This could have happened because the bolt

was defective or because of unexplained shearing of the bolt.

$ Bolt 4 shows a load-shelling profile similar to that

generated in the laboratory (see figure 3). The greatest load

was recorded 3.1 m above the back by gauges 3 and 8. The

ungrouted length (2.8 m) shows a low load primarily because

load was distributed over a large distance, as discussed

previously. It must be noted that loads were lower on the

gauges on either side of the bolt at 3.1 m. This implies that a

crack was close by.

$ Bolt 5 shows the classic load shelling profile. The highest

load was recorded in the ungrouted section within 2.8 m into

the back. The next highest load was at the bonded gauge 2.9 m

into the back. Readings again indicate that delamination was

largely controlled by cracks within 3 m into the back.

$ Loads on bolt 6 were uniform, with an average of 40 kN,

excepting for gauges 2 and 10, which were located 2.3 m above

the back. This implies that delamination may have been

occurring at this location. Note that this bolt was close to the

pillar boundary and therefore the depth of the fractured zone

would be less.
279

Figure 11.—Displacement of back over time.

$ Gauge sets 2 and 7 (2.3 m deep), 3 and 8 (2.5 m deep),

and 4 and 9 (2.7 m deep) on bolts 8 and 9 showed a load profile

approaching, and in most cases reaching, the yield strength of

the bolt. This reading indicates that delamination was occurring

throughout the horizon or that rock mass loading was higher.

These two bolts were placed in the north-center section of the

site, which experienced high loads such as those on bolt 4,

which had also been placed in the center of the intersection.

The deepest gauges (5 and 10) at 3.1 m into the back recorded

low values, indicating that the bond at depths between 2.7 and

3.1 m was sufficient to absorb most of the load.

$ Bolt 11 was located at a pillar boundary and therefore was

receiving more support from the mine rock, resulting in less

delamination and lower loads. Representative load curves for

the 3.1-m-deep gauge show the greatest amount of load. The

second set of gauges, at 2.3 m deep, took the next highest load,

and the first set of gauges, at 2.59 m deep, took the lowest

loads. Delamination was occurring, albeit at low loads,

approximately 3 m above the back.

$ Typical axial load curves are shown for bolt 201. These

types of curves usually result when conducting laboratory pull

tests on instrumented bolts. The curves also show a quick

loading rate, which should be expected because the bolt was

closest to active mining on the east side of the intersection.

$ Strain gauges on bolt 202 were reading both positive and

negative strain, which indicate bending of the bolt. Gauges 4

and 9 and gauges 5 and 10 indicated that bolt bending on one

side was in the yield zone of that bolt. Loads on gauge 5

showed the same loading horizon as the cable bolts at 2 to 5 m.

The remainder of the gauges showed that the nonbending

sections of the bolt were under low loads compared to the

sections of the bolt showing bending.

$ The tip on bolt 203 was undergoing high axial loads in the

yield zone. This could be interpreted as an indication that the

bolt was just reaching the load horizon at 2 to 3 m into the back.

The rest of the gauges were being loaded at relatively low load

rates.

$ The load curve on bolt 204 was typical except for

readings from two gauges. The bolt was supporting ground at

all instrumented horizons, with somewhat higher loads at the

center of the bolt. These load profiles also indicate that the

resin grout was developing a good bond along the entire length

of the hole. It is possible that a crack in the back caused grout

delamination that affected gauge set 3 and 7. Total lower load

readings along the bolt could have resulted from a greater bolt

density in the surrounding area.

RESULTS OF MEASUREMENTS 

AT SAGMETER STATIONS

Three sagmeter stations were installed at three different

depths (2.4, 3.2, and 3.8 m) at the intersection of 350S and

12,371E (figure 7). Figure 11 shows that sagmeters at the three

stations generated similar curves with displacements ranging

from 0.5 to 2.0 cm. In general, the most centrally located

station had the greatest displacement at all three depths.

Generally, the north-northwest and south-southeast sagmeter

stations, which were near the boundary pillars, showed the least

amount of deformation. Delamination was observed throughout

all horizons with the largest amount of movement between 2.4

and 3.8 m of the back. This roof movement corresponded with

data collected from the strain gauges, showing that

delamination was generally 2.4 m above the back. Sagmeter

readings from the 3.2-m-deep hole at the center of the

intersection would be 

comparable to readings from a 4.3-m-long cable bolt with

1.5 m of resin embedment. This would mean 2.8 m of

unbonded cable length.
280

The above case history is being further evaluated with

respect to deformation, mining sequence, and behavior of the

back. The depth of failure was most likely 3 m. Employing a

support pattern of bolts placed on 1.2- by 1.2-m spacings and a

specific gravity of 2.6 for the marlstone-trona-shale back would

yield a dead load of approximately 110 kN, which is within the

range of loads as measured by the bolts. Although several of

the measured loads exceeded the calculated dead load

significantly, which could have been the result of mininginduced loading, readings from these cable bolts enable

researchers to understand overall support-rock interaction.

CONCLUSIONS

In all installations where the gauges were read, cable bolts

provided information about loads in the immediate rock mass.

Grout quality could be determined by relating bonding capacity

of these cable bolts to the load profile of the ungrouted bolts.

The instrumented cable bolt was found to be reliable in

indicating load changes and has been correlated to deformation

of the adjacent rock mass.

The dangers of assuming that deformation is linear across a

bonded cable were also suggested as the load profile showed

the relationship of the degree of grout bonding to load along the

cable. The simple relationship of strain-to-load calculated with

cable modulus is only valid for the ungrouted sections of cable.

However, the grouted sections of the cable should be calculated

using calibrated data to determine the strain relationship. The

above interpretation is only possible if point strains and loads

on a cable are known.

This instrumented bolting system is used to aid in determining proper bolt spacings, bolt lengths, and load-carrying

characteristics of the bolts, thus allowing a mine manager to

design a safer roof support system for miners.

ACKNOWLEDGMENTS

We thank Tg Soda Ash and mine engineers John

McDonnell, Christopher Pritchard, and Scott Lehman; without

their cooperation, this research would not have been conducted.

We also thank Stephen P. Signer, mining engineer at SRL, for

his 

expertise in the application of Campbell Scientific systems; and

Richard Curtin, Dennis J. Cox, Richard L. Rains and Joanne L.

Johnston, engineering technicians at SRL, for their dedication

in constructing the instrumented bolt systems.

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