SECOND EDITION

Residential 

Structural 

Design

Guide

A State-of-the-Art Engineering 

Resource for Light-Frame Homes, 

Apartments, and Townhouses

U.S. Department of Housing and Urban Development | Office of Policy Development and Research 
Residential Structural Design Guide

A State-of-the-Art Engineering Resource 

for Light-Frame Homes,

Apartments, and Townhouses

Second Edition

Prepared for

U.S. Department of Housing and Urban Development

Office of Policy Development and Research

Prepared by

Coulbourne Consulting

October 2017
Residential Structural Design Guide ii

Acknowledgments

This document was prepared by Coulbourne Consulting, a sole proprietor structural engineering 

consultant, and sponsored by the U.S. Department of Housing and Urban Development. The 

principal authors of the guide are William L. Coulbourne, P.E., William Bracken, P.E., Dr. John 

van de Lindt, and Dr. David Prevatt, P.E. Graphics were produced by Bracken Engineering, and 

document editing and formatting was performed by Casey Thayer of Outreach Process Partners, 

LLC. Special appreciation is extended to Jay Crandell, P.E., Eric Stafford, P.E. and Dr. Daniel 

Dolan for their review and guidance throughout the project. Appreciation is also extended to the 

following individuals, whose comments made this work more complete and accurate: James 

Baty (American Concrete Institute), Fred Bowen (National Council of Structural Engineers 

Associations), Charles Carter (American Institute of Steel Construction), Kelly Cobeen (Wiss, 

Janney, Elstner Associates, Inc.), Dennis Graber (National Concrete Masonry Association), Gary 

Ehrlich (National Association of Home Builders), John Ingargiola (Federal Emergency 

Management Agency), Vladimir Kochkin (National Association of Home Builders Research 

Center), Dan Lavrich (American Society of Civil Engineers, Structural Engineering Institute), 

Phil Line (American Wood Council), Hamid Naderi (International Code Council), Larry Novak 

(Portland Cement Association), Long Phan (National Institute of Standards and Technology), 

Tim Reinhold (Insurance Institute for Business & Home Safety), Robert Ross (Forest Products 

Laboratory), Tim Smail (Federal Alliance for Safe Homes), and B.J. Yey (APA–The Engineered 

Wood Association). 

Disclaimer

Although the information in this document is believed to be accurate, neither the authors, the reviewers, the U.S. 

Department of Housing and Urban Development, Coulbourne Consulting, nor any of their employees or representatives 

makes any warranty, guarantee, or representation, expressed or implied, with respect to the accuracy, effectiveness, or 

usefulness of any information, method, or material in this document, nor assumes any liability for the use of any 

information, methods, or materials disclosed herein, or for damages arising from such use. This publication is intended 

for the use of professional personnel who are competent to evaluate the significance and limitations of the reported 

information and who will accept responsibility for the application of the material it contains. All responsibility as to the 

appropriate use of information in this document is the responsibility of the reader or user.

Residential Structural Design Guide iv

Preface

This document is a unique and comprehensive tool for design professionals, 

particularly structural engineers, seeking to provide value-added services to the producers 

and consumers of American housing. As such, the guide is organized around the 

following major objectives.

• To present a sound perspective on American housing relative to its history, 

construction characteristics, regulation, and performance experience.

• To provide the latest technical knowledge and engineering approaches for the 

design of homes to complement current code-prescribed design methods.

• To assemble relevant design data and methods in a single, comprehensive 

format that is instructional and simple to apply for the complete design of a 

home.

• To reveal areas in which gaps in existing research, design specifications, and 

analytic tools necessitate alternative methods of design and sound engineering 

judgment to produce efficient designs.

This guide consists of seven chapters. The layout and application of the various 

chapters are illustrated in the figure on page vi. Chapter 1 describes the basic substance of 

American housing, including conventional construction practices, alternative materials, 

building codes and standards, the role of design professionals, and actual experience with 

respect to performance problems and successes. Chapter 2 introduces basic engineering 

concepts regarding safety, load path, and the structural system response of residential 

buildings, subassemblies, and components to various types of loads. Chapter 3 addresses 

design loads applicable to residential construction. Chapters 4 and 5 provide step-by-step 

design procedures for the various components and assemblies comprising the structure of 

a home—from the foundation to the roof. Chapter 6 is devoted to the design of lightframe homes to resist lateral loads from wind and earthquakes. Chapter 7 addresses the 

design of various types of connections in a wood-framed home that are important to the 

overall function of the numerous component parts. As appropriate, the guide offers 

additional resources and references for the topics addressed.

Given that most homes in the United States are built with wood structural 

materials, the guide focuses on appropriate methods of design associated with wood for 

the above-grade portion of the structure. Concrete and masonry are generally assumed to 

be used for the below-grade portion of the structure, although preservative-treated wood 

may also be used. Other materials and systems using various innovative approaches are 

considered in abbreviated form, as appropriate. In some cases, innovative materials or 

systems can be used to address specific issues in the design and performance of homes. 

For example, cold-formed steel framing is becoming popular in many states because of 
Residential Structural Design Guide v

either the high cost of wood or concerns with decay and termite damage. Likewise, 

partially reinforced masonry construction is used extensively in Florida and the southeast 

because of its demonstrated ability to perform in high winds.

For typical wood-framed homes, the primary markets for engineering services lie 

in special load conditions, such as girder design for a custom house; corrective measures, 

such as repair of a damaged roof truss or floor joist; high-hazard conditions created by 

earthquakes on the West Coast, the intermountain west, the central United States, and 

South Carolina and by hurricanes along the Gulf and Atlantic coasts; and the increasing 

requirements for braced wall design in other locations. The design recommendations in 

this guide are based on the best information available to the authors for the safe and 

efficient design of homes. Much of the technical information and guidance is 

supplemental to building codes, standards, and design specifications that define current 

engineering practice. In fact, current building codes may not explicitly recognize some of 

the technical information or design methods described or recommended in the guide 

because the codes are minimums designed to protect lives, not to reduce property 

damage. A competent professional designer, therefore, should first compare and 

understand any differences between the content of this guide and local building code 

requirements. Any actual use of this guide by a competent professional may require 

appropriate substantiation as an “alternative method of analysis.” The guide and 

references provided herein should help furnish the necessary documentation.

Use of alternative means and methods of design should not be taken lightly or 

without first carefully considering the wide range of implications related to the applicable 

building code’s minimum requirements for structural design, the local process of 

accepting alternative designs, the acceptability of the proposed alternative design method 

or data, and exposure to liability when attempting something new or innovative, even 

when the process is carried out correctly. The intent of this guide is not to steer a designer 

unwittingly into noncompliance with current regulatory requirements for the practice of 

design as governed by local building codes. Instead, the intent is to provide technical 

insights into and approaches to home design that have not been compiled elsewhere but 

that deserve recognition and consideration. The guide is also intended to be instructional 

in a manner relevant to the current state of the art of home design.

Finally, it is hoped that this guide will foster a better understanding among 

engineers, architects, building code officials, and home builders by clarifying the 

perception of homes as structural systems. As such, the guide should help structural 

designers perform their services more effectively and assist in integrating their skills with 

other professionals who contribute to the production of safe and affordable homes in the 

United States.
Residential Structural Design Guide vi
Residential Structural Design Guide vii

Contents

Page 

1.1 Conventional Residential Construction ..................................................................... 1-1 

1.2 Factory-Built Housing ............................................................................................... 1-7 

1.3 Alternative Materials and Methods ............................................................................ 1-7 

1.4 Building Codes and Standards ................................................................................. 1-13 

1.5 Role of the Design Professional ............................................................................... 1-14 

1.6 Housing Structural Performance .............................................................................. 1-15 

1.7 Summary .................................................................................................................. 1-17 

1.8 References ................................................................................................................ 1-19 

 2.1 General ....................................................................................................................... 2-1 

 2.2 What Is Structural Design? ........................................................................................ 2-1 

 2.3 Load Types and Whole Building Response ............................................................... 2-3 

 2.4 Load Path ................................................................................................................... 2-7 

 2.5 Structural Reliability ................................................................................................ 2-15 

 2.6 References ................................................................................................................ 2-24 

 3.1 General ....................................................................................................................... 3-1 

 3.2 Load Combinations .................................................................................................... 3-2 

 3.3 Dead Loads ................................................................................................................ 3-4 

 3.4 Live Loads ................................................................................................................. 3-6 

 3.5 Soil Lateral Loads ...................................................................................................... 3-8 

 3.6 Wind Loads .............................................................................................................. 3-12 

 3.7 Snow Loads .............................................................................................................. 3-22 

 3.8 Earthquake Loads..................................................................................................... 3-25 

 3.9 Flood Loads ............................................................................................................. 3-32 

 3.10 Tornadoes ................................................................................................................. 3-33 

 3.11 Other Load Conditions ............................................................................................. 3-34 

 3.12 Design Examples ..................................................................................................... 3-36 

 3.13 References ................................................................................................................ 3-42 

Chapter 1—Basics of Residential Construction

Chapter 2—Structural Design Concepts

Chapter 3—Design Loads for Residential Buildings
Residential Structural Design Guide viii

 4.1 General ....................................................................................................................... 4-1 

 4.2 Material Properties ..................................................................................................... 4-5 

 4.3 Soil-Bearing Capacity and Footing Size .................................................................... 4-9 

 4.4 Footings.................................................................................................................... 4-12 

4.5 Foundation Walls ..................................................................................................... 4-23 

4.6 Slabs on Grade ......................................................................................................... 4-54 

4.7 Pile Foundations....................................................................................................... 4-56 

 4.8 Frost Protection ........................................................................................................ 4-58 

 4.9 Design Examples ..................................................................................................... 4-64 

 4.10 References ................................................................................................................ 4-88 

 5.1 General ....................................................................................................................... 5-1 

 5.2 Material Properties ..................................................................................................... 5-3 

 5.3 Structural Evaluation ............................................................................................... 5-14 

 5.4 Floor Framing .......................................................................................................... 5-23 

 5.5 Wall Framing ........................................................................................................... 5-33 

 5.6 Roofs ........................................................................................................................ 5-40 

 5.7 Design Examples ..................................................................................................... 5-50 

 5.8 References ................................................................................................................ 5-73 

 6.1 General ....................................................................................................................... 6-1 

 6.2 Overview of Whole-Building Tests ........................................................................... 6-3 

 6.3 LFRS Design Steps and Terminology ....................................................................... 6-7 

 6.4 The Current LFRS Design Practice ......................................................................... 6-13 

 6.5 Design Guidelines .................................................................................................... 6-21 

 6.6 Design Examples ..................................................................................................... 6-42 

 6.7 References ................................................................................................................ 6-68 

 7.1 General ....................................................................................................................... 7-1 

 7.2 Types of Mechanical Fasteners .................................................................................. 7-2 

 7.3 Wood Connection Design ........................................................................................ 7-11 

 7.4 Design of Concrete and Masonry Connections ....................................................... 7-23 

 7.5 Design Examples ..................................................................................................... 7-30 

 7.6 References ................................................................................................................ 7-45 

Appendix A—Shear and Moment Diagrams and Beam Equations

Appendix B—Unit Conversions

Chapter 5—Design of Wood Framing

Chapter 7—Connections

Chapter 6—Lateral Resistance to Wind and Earthquakes

Chapter 4—Design of Foundations
Residential Structural Design Guide ix

List of Figures

Page 

Figure 1.1a: Post-and-Beam Construction (Historical) ......................................................... 1-2 

Figure 1.1b: Balloon-Frame Construction (Historical) ......................................................... 1-3 

Figure 1.1c: Platform-Frame Construction (Modern) ........................................................... 1-4 

Figure 1.2: Modern Platform-Framed House Under Construction ..................................... 1-5 

Figure 1.3: House Construction Using Engineered Wood Components ............................. 1-9 

Figure 1.4: House Construction Using Cold-Formed Steel Framing ................................ 1-10 

Figure 1.5: House Construction Using Insulating Concrete Forms .................................. 1-11 

Figure 1.6: House Construction Using Concrete Masonry ............................................... 1-12 

Figure 2.1: Vertical Load Path for Gravity Loads ............................................................... 2-8 

Figure 2.2: Vertical Load Path for Wind Uplift .................................................................. 2-9 

Figure 2.3: Lateral Load Path ............................................................................................ 2-13 

Figure 2.4: Building Deformation Under Lateral Load ..................................................... 2-14 

Figure 2.5: Basic Concept of Safety in LRFD and ASD Considering 

 the Variability of Loads and Resistance ......................................................... 2-22 

Figure 3.1: Triangular Pressure Distribution on a Basement Foundation Wall .................. 3-9 

Figure 3.2a: Basic Design Wind Speed Map From ASCE 7-10 ......................................... 3-14 

Figure 3.2b: Basic Design Wind Speed Map From ASCE 7-10 ......................................... 3-15 

Figure 3.3a: Ground Snow Loads (ASCE 7-10) ................................................................. 3-23 

Figure 3.3b: Ground Snow Loads (ASCE 7-10) ................................................................. 3-24 

Figure 3.4a: Mapped Risk-Targeted Maximum Considered Earthquake (MCER) 

 Spectral Response Acceleration ...................................................................... 3-26 

Figure 3.4b: Mapped Risk-Targeted Maximum Considered Earthquake (MCER) 

 Spectral Response Acceleration ...................................................................... 3-27 

Figure 4.1: Types of Foundations ........................................................................................ 4-4 

Figure 4.2: Critical Failure Planes in Continuous or Square Concrete Spread Footings .. 4-16 

Figure 4.3: Shear Calculations for Plain Concrete Walls: Variables Defined .................. 4-26 

Figure 4.4: Shear Calculations in Reinforced Concrete Walls: Variables Defined .......... 4-29 

Figure 4.5: Typical Interaction Diagrams for Plain and Reinforced Concrete Walls ....... 4-33 

Figure 4.6: Design Variables Defined for Lintel Bending and Shear ............................... 4-36 

Chapter 1—Basics of Residential Construction

Chapter 2—Structural Design Concepts

Chapter 4—Design of Foundations

Chapter 3—Design Loads for Residential Buildings
Residential Structural Design Guide x

Figure 4.7: Shear Calculations in Reinforced Concrete 

Masonry Walls: Variables Defined ................................................................. 4-46 

Figure 4.8: Concrete Masonry Wall Lintel Types ............................................................. 4-49 

Figure 4.9: Preservative-Treated Wood Foundation Walls ............................................... 4-51 

Figure 4.10: Insulating Concrete Form Foundation Walls .................................................. 4-54 

Figure 4.11: Basic Coastal Foundation Construction .......................................................... 4-57 

Figure 4.12: Air-Freezing Index Map (100-Year Return Period) ....................................... 4-61 

Figure 4.13: Frost-Protected Shallow Foundation Applications ......................................... 4-63 

Figure 5.1: Components and Assemblies of a Conventional Wood-Framed Home ........... 5-2 

Figure 5.2: Structural Elements of the Floor System ........................................................ 5-25 

Figure 5.3: Conventional and Alternative Floor Framing Members ................................. 5-27 

Figure 5.4: Examples of Beams and Girders ..................................................................... 5-30 

Figure 5.5: Structural Elements of the Wall System ......................................................... 5-34 

Figure 5.6: Wood Column Types ...................................................................................... 5-39 

Figure 5.7: Structural Elements of a Conventional Roof System ..................................... 5-41 

Figure 5.8: Design Methods and Assumptions for a Sloped Roof Rafter ......................... 5-44 

Figure 5.9: Typical Roof Overhang Constructions ........................................................... 5-48 

Figure 5.10: Gable-End Wall Bracing ................................................................................. 5-50 

Figure 6.1: Twin Garages on Shake Tables, SUNY Buffalo .............................................. 6-5 

Figure 6.2: Six-Story Building on Shake Table, Miki, Japan ............................................. 6-6 

Figure 6.3: Chords in Shear Walls and Horizontal Diaphragms Using the 

 “Deep Beam” Analogy ..................................................................................... 6-9 

Figure 6.4: Shear Wall Collector and Composite Failure Plane ....................................... 6-10 

Figure 6.5: Two Types of Hold-Down Restraint and Basic Analytic Concepts ............... 6-12 

Figure 6.6: Lateral Force Distribution by a “Flexible” Diaphragm 

 (Tributary Area Approach) ............................................................................. 6-14 

Figure 6.7: Basic Perforated Shear Wall ........................................................................... 6-19 

Figure 6.8: Evaluation of Overturning Forces on a Restrained Shear Wall Segment ....... 6-33 

Figure 6.9: Moment Frame ................................................................................................ 6-38 

Figure 7.1: Elements of a Nail and Nail Types ................................................................... 7-3 

Figure 7.2: Bolt and Connection Types ............................................................................... 7-7 

Figure 7.3: Specialty Connector Hardware ....................................................................... 7-10 

Figure 7.4: Types of Connections and Loading Conditions .............................................. 7-13 

Figure 7.5: Concrete or Masonry Wall-to-Footing Connections ....................................... 7-25 

Figure 7.6: Key in Concrete Footings ............................................................................... 7-26 

Figure 7.7: Dowel Placement in Concrete Footings .......................................................... 7-27

Chapter 5—Design of Wood Framing

Chapter 6—Lateral Resistance to Wind and Earthquakes

Chapter 7—Connections
Residential Structural Design Guide xi

Appendix A—Shear and Moment Diagrams and Beam Equations

Figure A.1: Simple Beam (Foundation Wall)—Partial Triangular Load ............................ A-1 

Figure A.2: Simple Beam (Wall or Column)—Eccentric Point Load ................................ A-2 

Figure A.3: Simple Beam—Uniformly Distributed Load ................................................... A-2 

Figure A.4: Simple Beam—Load Increasing Uniformly to One End ................................. A-3 

Figure A.5: Simple Beam—Concentrated Load at Any Point ............................................ A-3 

Figure A.6: Simple Beam—Two Unequal Concentrated Loads Unsymmetrically Placed A-4 

Figure A.7: Cantilever Beam—Uniformly Distributed Load ............................................. A-4 

Figure A.8: Cantilever Beam—Concentrated Load at Any Point ....................................... A-5 

Figure A.9: Beam Fixed at One End, Supported at Other—Uniformly Distributed Load .. A-5 

Figure A.10: Beam Fixed at One End, Supported at Other—Load at Any Point ................. A-6 

Figure A.11: Beam Fixed at Both Ends—Uniformly Distributed Load ............................... A-6 

Figure A.12: Beam Fixed at Both Ends—Concentrated Load at Any Point ......................... A-7 

Figure A.13: Beam Overhanging One Support—Uniformly Distributed Load .................... A-7 

Figure A.14: Beam Overhanging One Support—Concentrated Load at End of Overhang .. A-8 

Figure A.15: Continuous Beam—Two Equal Spans and Uniformly Distributed Load ........ A-8 

Figure A.16: Continuous Beam—Two Equal Spans With Uniform Load on One Span ...... A-9 

Figure A.17: Continuous Beam—Two Unequal Spans and Uniformly Distributed Load ... A-9 

 
Residential Structural Design Guide xii

List of Tables

Page 

Table 2.1: Building Loads Categorized by Orientation 2-3 

Table 3.1: Typical Load Combinations Used for the Design of Components 

and Systems 3-4 

Table 3.2: Dead Loads for Common Residential Construction 3-5 

Table 3.3: Densities for Common Residential Construction Materials 3-6 

Table 3.4: Live Loads for Residential Construction 3-7 

Table 3.5: Live Load Element Factor, KLL 3-8 

Table 3.6: Values of Ka, Soil Unit Weight, and Equivalent Fluid Density 

 by Soil Type 3-10 

Table 3.7: Lateral Soil Load 3-11 

Table 3.8: Basic Wind Velocity Pressures (psf) for Suburban Terrain (MWFRS) 3-16 

Table 3.9: Lateral Pressure Coefficients for Application to Vertical Projected Areas 3-17 

Table 3.10: Wind Pressure Coefficients for Systems and Components 3-19 

Table 3.11: Missile Types for Wind-Borne Debris Impact Tests 3-21 

Table 3.12: Site Soil Amplification Factor Relative to Acceleration

 (short period, Site Class D) 3-29 

Table 3.13: Seismic Design Factors for Residential Construction 3-30 

Table 4.1: Rebar Size, Diameter, and Cross-Sectional Areas 4-7 

Table 4.2: Presumptive Soil-Bearing Values by Soil Description 4-10 

Table 4.3: Presumptive Soil-Bearing Values (psf) Based on Standard 

 Penetrometer Blow Count 4-11 

Table 4.4: Minimum Compressive Strength f’c at 28 days and Maximum 4-24 

 Slump of Concrete 

Table 4.5: Simplified Moment Magnification Factors, δns 4-32 

Table 4.6: Nominal Wall Thickness for 8-Foot High Masonry Foundation Walls 4-39 

Table 4.7: Allowable Flexural Tension Stresses (Fa) for Allowable Stress Design 

 of Unreinforced Masonry 4-40 

Table 4.8: Preservative-Treated Wood Foundation Framing 4-52 

Table 4.9: Minimum Frost Depths for Residential Footings 4-59 

Chapter 3—Design Loads for Residential Buildings

Chapter 2—Structural Design Concepts

Chapter 4—Design of Foundations
Residential Structural Design Guide xiii

Table 5.1: Adjustment Factor Applicability to Design Values for Wood 5-10 

Table 5.2: Recommended Load Duration Factors for ASD 5-12 

Table 5.3: Recommended Repetitive Member Factors for Dimension

 Lumber Used in Framing Systems 5-13 

Table 5.4: Recommended Allowable Deflection Limits 5-20 

Table 5.5: Fastening Floor Sheathing to Structural Members 5-32 

Table 6.1: Nominal Shear Resistance (plf) for Wood Structural Panel Shear Walls 

with Framing of Douglas-Fir, Larch, or Southern Pine 6-24 

Table 6.2: Nominal Unit Shear Resistance (plf) for Walls with 

Cold-Formed Steel Framing and Wood Structural Panels 6-25 

Table 6.3: Nominal Unit Shear Values (plf) for 1/2-Inch Thick Gypsum 

Wall-board Sheathing 6-26 

Table 6.4: Horizontal Diaphragm Maximum Nominal Shear Values (plf) for 

 Unblocked Roof and Floor Construction Using Douglas-Fir Framing 6-39 

Table 7.1: Nail Types, Sizes, and Dimensions 7-5 

Table 7.2: Common Framing Lumber Species and Specific Gravity Values 7-12 

Chapter 5—Design of Wood Framing

Chapter 6—Lateral Resistance to Wind and Earthquakes

Chapter 7—Connections
Residential Structural Design Guide 1-1

CHAPTER 1

Basics of Residential 

Construction

1.1 Conventional Residential 

Construction

The conventional American house has been shaped over time by a variety 

of factors. Foremost, the abundance of wood as a readily available resource has 

dictated traditional American housing construction, beginning as log cabins, then 

as post-and-beam structures, and finally as light-frame buildings. The basic 

residential construction technique, which has remained much the same since the 

introduction of light wood-framed construction in the mid-1800s, is generally 

referred to as conventional construction. See figures 1.1a through 1.1c for 

illustrations of various historical and modern construction methods using wood.

Today, a wood framed residential building can be typically constructed in one of 

two ways: (1) conventionally framed, constructed from wall panels built in a 

factory, and assembled on the jobsite, or (2) built in a factory and brought to a 

jobsite and placed on a site-built foundation.

In post-and-beam framing, structural columns support horizontal 

members. Post-and-beam framing is typified by the use of large timber members. 

Traditional balloon framing consists of closely spaced light vertical structural 

members that extend from the foundation sill to the roof plates. Platform framing

is the modern adaptation of balloon framing, whereby vertical members extend 

from the floor to the ceiling of each story. Balloon and platform framings are not 

simple adaptations of post-and-beam framing but are actually unique forms of 
Residential Structural Design Guide 1-2

wood construction. Platform framing is used today in most wood-framed 

buildings; however, variations of balloon framing may be used in certain parts of 

otherwise platform-framed buildings, such as great rooms, stairwells, and gableend walls, where continuous wall framing provides greater structural integrity. 

Figure 1.2 depicts a modern home under construction.

FIGURE 1.1a Post-and-Beam Construction (Historical)
Residential Structural Design Guide 1-3

FIGURE 1.1b Balloon-Frame Construction (Historical)
Residential Structural Design Guide 1-4

FIGURE 1.1c Platform-Frame Construction (Modern)
Residential Structural Design Guide 1-5

FIGURE 1.2 Modern Platform-Framed House Under Construction

Conventional or prescriptive construction practices are based as much on 

experience as on technical analysis and theory (HEW, 1931). The minimum 

building code requirements provided by the International Residential Code (IRC) 

have codified conventional construction practices but do have some basis in basic 

engineering principles. The prescriptive construction requirements provided in the 

IRC are intended to be easy for a builder to follow and for a code official to 

inspect without the services of a design professional. It is also common for design 

professionals, including architects and engineers, to apply conventional practices 

in typical design conditions but to undertake special designs for certain parts of a 

home that are beyond the scope of the IRC or a prescriptive residential design 

guide. It is very important for design professionals to understand the limitations of 

the prescriptive code when relying on it.. The housing market historically has 

operated with minimal involvement of design professionals. As building codes 

advance, environmental loads become better understood, and performance 

demands on residential construction continue to increase, however, so too does 

the role of the design professional. Section 1.5 explores the current role of design 

professionals in residential construction.
Residential Structural Design Guide 1-6

Although dimensional lumber has remained the predominant material used 

in the last century of American housing construction, the size of the material has 

been reduced from the rough-sawn, 2-inch-thick members used in the late 1800s 

to today’s nominal dressed sizes, with actual thicknesses of less than 1.5 inches 

for standard framing lumber. The result has been a significant improvement in 

economy and resource use accompanied by significant structural tradeoffs. 

The mid-to-late 1900s also saw several significant innovations in preengineered wood products and wood-framed construction techniques. One 

example is the development of the metal plate-connected wood truss in the 1950s. 

Metal plate-connected wood trusses, most often referred to as pre-engineered 

wood trusses, are now used in many new homes because the pre-engineered 

method is generally more efficient than older framing methods that rely on roof 

rafters. In addition to being used in roof framing, pre-engineered wood trusses and 

beams are also used in floor framing. As floor framing, these trusses are able to 

increase floor rigidity and the spans of flooring systems, eliminating some interior 

load-bearing walls. Other examples of innovative products and techniques are 

plywood structural sheathing panels that entered the market in the 1950s and 

oriented strand board (OSB) that entered the market in the 1980s. Both products 

quickly replaced board sheathing on walls, floors, and roofs. 

It is important to recognize that, while the previously mentioned changes 

in materials and methods were occurring, significant changes in house designs 

also occurred, in the way of larger homes with more complicated architectural 

features, long-span floors and roofs, and large open interior spaces. The collective 

effect of these changes on the structural qualities of most homes is certainly 

notable.

The following references are recommended for a more in depth 

understanding of conventional housing design, detailing, and construction.

Section 1.8—References—provides detailed citations.

• 2012 International Residential Code (ICC, 2012a).

• 2012 International Building Code (ICC, 2012b).

• Wood Frame Construction Manual (AWC, 2012b).

• Modern Carpentry—Building Construction Details in Easy-toUnderstand Form, 10th ed. (Wagner, 2003).

The following structural design references are also recommended for use 

with chapters 3 through 7 of this guide.

• NDS—National Design Specification for Wood Construction and 

Supplement (AWC, 2012a).

• ACI-318—Building Code Requirements for Structural Concrete (ACI, 

2011a).

• ACI-530—Building Code Requirements for Masonry Structures (ACI, 

2011).

• ASCE 7-10—Minimum Design Loads for Buildings and Other

Structures (ASCE, 2010).
Residential Structural Design Guide 1-7

1.2 Factory-Built Housing

Most homes in the United States are still site built; that is, they follow a 

stick-framing approach. With this method, wood members are assembled on site 

from the foundation up. The primary advantage of onsite building is flexibility in 

meeting variations in housing styles, design details, and changes specified by the 

owner or builder. An increasing number of today’s site-built homes, however, use 

components that are fabricated in an offsite plant. Prime examples include wall 

panels (both structural insulated panels [SIPs] and those built with dimensional 

lumber) and pre-engineered wood trusses. The blend of stick-framing and factorybuilt components is referred to as component building.

Modular housing is a step beyond component building. Modular housing

is constructed in essentially the same manner as site-built housing except that 

houses are factory-built in finished modules (typically two or more modules) and 

shipped to the jobsite for placement on site-built foundations. Modular housing is 

built to comply with the same building codes that govern site-built housing. 

Manufactured housing (formerly known as mobile homes) is also 

constructed using wood-framed methods and components; however, these 

methods and components are required to comply only with the federal preemptive 

standards specified in the Manufactured Housing Construction Safety Standards 

(U.S. Department of Housing and Urban Development code). This popular form 

of industrialized housing is completely factory assembled and then delivered to a 

site by using an integral chassis for over-the-road travel and foundation support. 

1.3 Alternative Materials and Methods 

Several innovations in structural materials have been introduced more 

recently to residential construction. Alternatives to conventional wood-framed 

construction are in fact gaining recognition in modern building codes. It is important 

for designers to become familiar with these alternatives because their effective 

integration into conventional home building may require the services of a design 

professional. In addition, a standard practice in one region of the country may be 

viewed as an alternative in another, which provides opportunities for innovation 

across regional norms.

Many options in the realm of materials are already available. The 

following pages describe several significant examples. In addition, the following 

contacts are useful for obtaining design and construction information on the 

alternative materials and methods for house construction.

General contacts

HUD User (http://huduser.gov).

ToolBase (http://toolbase.org).

Engineered wood products

American Wood Council (http://awc.org).
Residential Structural Design Guide 1-8

APA–The Engineered Wood Association (http://apawood.org).

Structural Building Components Association (http://sbcindustry.com).

Cold-formed steel

Steel Framing Alliance (http://steelframingalliance.com).

American Iron and Steel Institute (http://steel.org).

Cold-Formed Steel Engineers Institute (http://cfsei.org).

Insulating concrete forms

EPS Industry Alliance (http://forms.org).

Structural Insulated Panels

Structural Insulated Panel Association (http://www.sips.org).

Masonry

National Concrete Masonry Association (http://ncma.org).

Engineered wood products and components (see figure 1.3) have 

gained considerable popularity in the past 30 years. Engineered wood products 

and components include wood-based materials and assemblies of wood products 

with structural properties similar to or better than the sum of their component

parts. Examples include metal plate-connected wood trusses, wood I-joists, 

laminated veneer lumber (LVL), plywood, oriented strand board (OSB), gluelaminated lumber, and parallel strand lumber (PSL). OSB structural panels are 

rapidly displacing plywood as a favored product for wall, floor, and roof 

sheathing. Wood I-joists are now used in 54 percent of the total framed floor area 

in all new homes each year (APA, 2013). Cross-laminated timber, (CLT) is now 

being manufactured in Canada, consists of laminated layers of solid sawn or 

structural composite lumber that are bonded with structural adhesives to form a 

rectangular-shaped timber. This product is expected to be more widely available 

in the United States in coming years.

The increased use of engineered wood products is the result of many years 

of research and product development and, more importantly, reflects the 

economics of the building materials market. Engineered wood products often 

offer improved dimensional stability, increased structural capability, ease of 

construction, and more efficient use of the nation’s lumber resources, and they do 

not require a significant change in construction technique. 

The designer should, however, carefully consider the unique detailing and 

connection requirements associated with engineered wood products and ensure 

that the requirements are clearly understood in the design office and at the jobsite. 

Design guidance, such as span tables and construction details, is usually available 

from the manufacturers of these predominantly proprietary products. A note of 

caution: for these proprietary products to be supported by the manufacturer, they 

must be installed exactly in accordance with the manufacturer’s instructions.
Residential Structural Design Guide 1-9

FIGURE 1.3 House Construction Using Engineered Wood Components

Cold-formed steel framing (previously known as light-gauge steel 

framing) was originally produced by a fragmented industry with nonstandardized 

products serving primarily the commercial design and construction market. In 

cooperation with the industry, HUD sponsored research necessary to develop 

standard minimum dimensions and structural properties for basic cold-formed 

steel framing materials, which resulted in the development of IRC design 

provisions. Cold-formed steel framing is currently used in exterior and interior 

walls of new housing starts. The benefits of cold-formed steel include low cost, 

durability, light weight, and strength (HUD, 1994). Figure 1.4 illustrates the use 

of cold-formed steel framing in a home. Construction methods can be found in the 

International Residential Code (ICC, 2012a). 
Residential Structural Design Guide 1-10

FIGURE 1.4 House Construction Using Cold-Formed Steel Framing 

Insulating concrete form (ICF) construction, as illustrated in figure 1.5, 

combines the forming and insulating functions of concrete construction in a single 

step. In a cooperative effort between the housing industry and HUD, the product 

class was included in the I-Codes after the establishment of minimum dimensions 

and standards for ICF construction. The benefits of ICF construction include 

durability, strength, noise control, and energy efficiency (HUD, 1998a; HUD, 

1998b). The method, detailed in Prescriptive Method for Insulating Concrete

Forms in Residential Construction, has been adopted by the IRC and is also 

discussed in the Prescriptive Design of Exterior Concrete Walls (PCA, 2012).
Residential Structural Design Guide 1-11

FIGURE 1.5 House Construction Using Insulating Concrete Forms

Structural insulated panels (SIPs), are composite panels of polystyrene 

or polyurethane foam sandwiched on both sides with OSB sheets. The panel size 

is typically the same as the manufactured size of the OSB sheets, but SIPs can 

also be larger. Individual SIPs are connected together by a vertical spline (splice) 

consisting of a 3-inch-wide OSB strip that bridges an expansion gap between the 

SIPs and is nailed to the OSB sheets on each side. A sufficient amount of foam is 

removed to allow the top and bottom plates to fit snugly inside the OSB. For 

additional stiffness, a further section of foam can be removed to accommodate 

abutting 2x studs or a foam block at the spline. This construction method 

eliminates the need for other insulation on the walls and roofs. 

Concrete masonry construction, illustrated in figure 1.6, has remained 

essentially unchanged in its basic construction method. Recently introduced 

products offer innovations that provide structural and architectural benefits, 

however. Masonry construction is well recognized for its fire-safety qualities, 

durability, noise control, termite resistance, and strength. The installed cost of 

masonry construction, like most alternatives to conventional wood-framed 

construction, may be a local issue that needs to be balanced against other factors.

For example, in hurricane-prone regions along the Gulf Coast and southern 

Atlantic states, standard concrete masonry construction dominates the market 
Residential Structural Design Guide 1-12

because its performance in major hurricanes has been favorable when nominally 

reinforced using conventional practices. 

FIGURE 1.6 House Construction Using Concrete Masonry

Reinforced concrete construction is a frequently used material and 

method in nonresidential construction that is gaining popularity in home 

construction in some parts of the country. This gain is because of its performance 

in extremely hot climates, in those locations with termite or woodboring insect 

issues, in those locations subject to either hurricane- or tornado-force winds, and 

for those building owners who want an exterior less prone to deterioration and 

severe weathering. Construction techniques in forming and pouring concrete for

homes are the same as used for nonresidential construction.

Alternative materials and methods provisions exist within the IRC and 

the International Building Code (IBC). These building code provisions provide 

the flexibility for a design professional or builder to use new materials in 

construction that may not be discussed or even contemplated in building codes. 

The IRC and IBC provide this flexibility within chapter 1, which describes a 

process whereby the designer or builder and the code official can review and 

approve such approaches.
Residential Structural Design Guide 1-13

1.4 Building Codes and Standards

Most of the U.S. population lives in areas that are covered by legally 

enforceable building codes that govern the design and construction of buildings, 

including residential dwellings. Although building codes are legally a 

governmental police power, most states allow local political jurisdictions to adopt 

or modify building codes to suit their special needs or, in a few cases, to write 

their own code. Almost all jurisdictions adopt a family of model codes by 

legislative action instead of attempting to write their own code.

The dominant family of model building codes in the United States is that 

developed by the International Code Council (ICC). The ICC was founded in 

1994 by the three regional code organizations—Building Officials and Code 

Administrators International, Inc.; International Conference of Building Officials; 

and Southern Building Code Congress International, Inc. This initiative was the 

result of the conclusion by the founders that the nation needed a single set of 

model building codes. The ICC has developed codes for all types of buildings and 

occupancies—from a backyard storage shed to a highrise office building and 

sports complex. In addition, some jurisdictions have also adopted building codes 

developed by the National Fire Protection Association (NFPA). The two major 

building code organizations are—

• International Code Council

500 New Jersey Avenue, NW

Washington, DC 20001

http://iccsafe.org

• National Fire Protection Association

1 Batterymarch Park

Quincy, MA 02169

http://nfpa.org

In the past, although the dominant codes included some “deemed-tocomply” prescriptive requirements for conventional house construction, they 

focused primarily on performance (that is, engineering) requirements. By 

focusing more on performance requirements, these codes were better able to 

address more complex buildings across the whole range of occupancy and 

construction types. Therefore, in an effort to provide a comprehensive, easier to 

use code for residential construction, the IRC was developed. Presented in logical 

construction sequence, the IRC is devoted entirely to simple prescriptive 

requirements for one- and two-family dwellings, duplexes, and townhouses. Many 

state and local jurisdictions have adopted both the IRC and the IBC. Thus, 

designers and builders enjoy a choice as to which set of requirements best suits 

their purpose.

Model building codes do not provide detailed specifications for all 

building materials and products but rather refer to established industry standards, 
Residential Structural Design Guide 1-14

such as those promulgated by ASTM International, formerly known as the 

American Society for Testing and Materials (ASTM). Several ASTM standards 

are devoted to the measurement, classification, and grading of wood properties for 

structural applications and of virtually all other building materials, including steel, 

concrete, and masonry. Design standards and guidelines for wood, steel, concrete, 

and other materials or applications are also maintained as reference standards in 

building codes. More than 600 materials and testing standards from a variety of 

organizations currently are referenced in the building codes used in the United 

States.

For products and processes not explicitly recognized in the codes or 

standards, the ICC Evaluation Service, Inc. (ICC-ES) provides evaluations of 

products relative to the model code requirements. The ICC-ES report recognizes a 

specific building product’s ability to meet the performance and prescriptive 

provisions in the code. It is an independent finding of the product’s capability. 

The report provides engineers the assurance of validity and technical accuracy in 

determining a product’s correct application. Reports are valid for a specific period 

of time. A report can undergo revisions at any time. Other organizations—such as 

Intertek, the International Association of Plumbing and Mechanical Officials, and 

Miami-Dade County, Florida—provide testing of building products for 

performance certifications and building code compliance.

Seasoned designers spend countless hours in careful study and application 

of building codes and selected standards that relate to their area of practice. These 

designers develop a sound understanding of the technical rationale and intent 

behind various provisions in applicable building codes and design standards. This 

experience and knowledge, however, can become even richer when coupled with 

practical experiences from the field. One of the most valuable sources of practical 

experience is the study of the successes and failures of past designs and 

construction practices, as presented in section 1.6.

1.5 Role of the Design Professional

Because the primary user of this guide is assumed to be a design 

professional, it is important to understand the role that design professionals can 

play in the residential construction process, particularly regarding recent trends. 

Design professionals offer a wide range of services to builders or developers in 

the areas of land development, environmental impact assessments, geotechnical 

and foundation engineering, architectural design, structural engineering, and 

construction monitoring. This guide, however, focuses on two approaches to 

design, as follows.

• Conventional design. Sometimes referred to as “prescriptive” 

construction, conventional design relies on standard practice and 

empirical methods as governed by prescriptive building code 

requirements (see section 1.4). This prescriptive approach, however, 

does not preclude and may even require some parts of the structure to 

be specially designed by an engineer or architect.
Residential Structural Design Guide 1-15

• Engineered design. Engineered design generally involves the 

application of engineering practice as represented within the building 

codes and design standards.

Some of the conditions that typically cause concern in the planning and 

preconstruction phases of home building and thus sometimes create the need for 

professional design services are—

• Structural configurations, such as unusually long floor spans, 

unsupported wall heights, large openings, or long-span cathedral 

ceilings.

• Loading conditions, such as high winds, high seismic risk, flood risks, 

coastal construction, heavy snows, or abnormal equipment loads.

• Engineering certifications, such as those required in V-zone flood 

areas and California seismic areas.

• Nonconventional building systems or materials, such as composite 

materials, structural steel, or unusual connections and fasteners.

• Geotechnical or site conditions, such as expansive soil, variable soil or 

rock foundation bearing, flood-prone areas, high water tables, or 

steeply sloped sites.

• Owner’s requirements, such as special materials, appliance or fixture 

loads, atriums, and other special features.

Although some larger production builders produce sufficient volume to 

justify employing a full-time design professional, most builders use consultants 

on an as-needed basis. As more and more homes are built in earthquake-prone 

areas and along the hurricane-prone coastlines, however, the involvement of 

structural design professionals is increasing. The added complexities of larger 

custom-built homes and special site conditions further serve to spur demand for 

design professionals. Moreover, if nonconventional materials and methods of 

construction are to be used effectively, the services of a design professional are 

often required. In some instances, builders in high-hazard areas are using design 

professionals for onsite compliance inspection in addition to building design.

1.6 Housing Structural Performance

1.6.1 General

Of the more than 130 million housing units in the United States, 

approximately two-thirds are single-family dwellings. With that many units in 

service, a substantial number can be expected to experience performance 

problems, most of which amount to minor defects that are easily detected and 

repaired. Other performance problems, such as foundation problems related to 

subsurface soil conditions, are unforeseen or undetected and may not be realized 

for several years.
Residential Structural Design Guide 1-16

On a national scale, tens of thousands of homes are subjected to extreme 

climatic or geologic events in any given year. Of that number, some will be 

damaged because of events that exceed the performance expectations of the 

building code (that is, a direct tornado strike or a large-magnitude hurricane, 

thunderstorm, flood, or earthquake). In addition, some will experience problems 

resulting from defective workmanship, premature product failure, design flaws, or 

durability problems (that is, rot, termites, or corrosion). Often, a combination of 

factors leads to the most dramatic forms of damage. Because the cause and effect 

of these problems do not usually fit simple generalizations, it is important to 

consider cause and effect objectively in terms of the overall housing inventory. 

The role of building codes historically has been to ensure that an 

acceptable level of safety is maintained during the life of a house to limit lifethreatening performance problems. Because the public may not benefit from an 

excessive degree of safety, code requirements must also maintain a reasonable 

balance between affordability and safety. As implied by any rational 

interpretation of a building code or design objective, safety must include an 

accepted level of risk. In this sense, economy, energy efficiency, sustainability, 

and affordability may be broadly considered as competing performance 

requirements. For a designer, the challenge is to consider optimum value and to 

use cost-effective design methods that result in acceptable building performance 

in keeping with the intent of the building code. In many cases, designers may be 

able to offer cost-effective options to builders and owners that improve 

performance well beyond the expected norm. Owners, however, must understand 

that they carry the burden of risk beyond what is implied in the building built “to 

code”.

Building codes today are focusing more on life-cycle performance—

including durability, sustainability, energy usage, and efficiency—in addition to 

life safety. These building code requirements include improved performance in 

response to normal and common occurrences such as water leaks, sagging floors, 

surface resistance to weathering, and temperature extremes. Building code 

requirements also include improved performance in response to less frequent 

occurrences such as floods, hurricanes, and earthquakes. The designer needs to be 

familiar with techniques to improve performance in all these situations to better 

serve the client.

1.6.2 Common Building Performance Issues

Common building performance issues have been found to include water 

intrusion of building envelopes; water intrusion to basements and foundations; 

building movements because of soil conditions; and failures of roof coverings, 

exterior claddings, and interior finishes.

These issues do not result solely from building products, because builders 

are often averse to products that are “too new.” Products and systems that have 

been the subject of class-action lawsuits in the United States give builders some 

reason to think twice about specifying new products. Examples of such products 

and systems include—
Residential Structural Design Guide 1-17

• Exterior Insulation and Finish Systems (EIFS).

• Fire-retardant-treated, or FRT, plywood roof sheathing.

• Certain composite sidings and exterior finishes.

• Polybutylene water piping.

Recent issues with heavily used products that had been long accepted 

(formaldehyde in wood products and contaminated drywall) have served to 

reinforce builders’ concerns about product performance.

Note that many of these problems have been resolved by subsequent 

product improvements. It is unfortunately beyond the scope of this guide to give a 

complete account of the full range of problems experienced in housing 

construction.

1.6.3 Housing Performance in Hurricanes, Earthquakes, 

Floods, and Tornadoes

Scientifically designed studies of housing performance during natural 

disasters have permitted objective assessments of actual performance relative to 

that intended by building codes. Anecdotal damage studies, conversely, are often 

subject to notable bias. Both objective and subjective damage studies provide 

useful feedback to builders, designers, code officials, and others with an interest 

in housing performance. The issue of housing performance in high-hazard areas 

will continue to increase in importance, because nearly 50 percent of the U.S. 

population lives along coastlines, raising concerns about housing safety, 

affordability, and durability. Therefore, it is essential that housing performance be 

understood objectively as a prerequisite to guiding rational design and 

construction decisions. Proper design that takes into account the wind and 

earthquake loads discussed in chapter 3 and the structural analysis procedures 

addressed in chapters 4, 5, 6, and 7 will likely result in efficient designs that 

address the performance issues discussed in those chapters. Regardless of the 

efforts made in design, however, the intended performance can be realized only 

with an adequate emphasis on installed quality. For this reason, some builders in 

high-hazard areas have retained the services of design professionals as much for 

onsite compliance inspections as for their design services. This practice offers 

additional quality assurance to the builder, designer, and owner in high-hazard 

areas of the country. It is within these extreme events that most performance 

problems are observed, manifested, or exacerbated.

1.7 Summary

Housing in the United States has evolved over time under the influence of 

a variety of factors. Although available resources and the economy continue to 
Residential Structural Design Guide 1-18

play significant roles, building codes, consumer preferences, and alternative 

construction materials are becoming increasingly important factors. In particular, 

building codes in the United States require homes in many special high-hazard 

areas to be designed by design professionals rather than by following prescriptive 

construction practices. This apparent trend may be attributed in part to changing 

perceptions regarding housing performance in these high-risk areas. Therefore, 

greater emphasis must be placed on the efficient structural design of housing. 

Although efficient design should also strive to improve construction quality 

through simplified construction, it also places greater importance on the quality of 

installation required to achieve the intended performance without otherwise 

relying on overdesign to compensate partially for real or perceived problems in 

installation quality.
Residential Structural Design Guide 1-19

1.8 References

American Concrete Institute (ACI). 2011a. Building Code Requirements for 

Structural Concrete and Commentary. ACI Standard 318. Farmington 

Hills, MI: American Concrete Institute.

———. 2011b. Building Code Requirements and Specifications for Masonry 

Structures and Companion Commentaries. ACI Standard 530. Farmington 

Hills, MI: American Concrete Institute.

American Plywood Association (APA). 2013. Market Outlook and Regional 

Production. APA Economics Report E79, 2013–2017. Tacoma, WA: 

American Plywood Association.

American Society of Civil Engineers (ASCE). 2010. Minimum Design Loads for 

Buildings and Other Structures. ASCE Standard 7-10. Reston, VA: 

American Society of Civil Engineers.

American Wood Council (AWC). 2012a. National Design Specification for Wood 

Construction and Supplement. Leesburg, VA: American Wood Council.

——. 2012b. Wood Frame Construction Manual. Leesburg, VA: American Wood 

Council.

International Code Council (ICC). 2012a. International Residential Code, 2012 

ed. Washington, DC: International Code Council.

——. 2012b. International Building Code, 2012 ed. Washington, DC: 

International Code Council.

Portland Cement Association (PCA). 2012. Prescriptive Design of Exterior 

Concrete Walls. Skokie, IL: Portland Cement Association.

U.S. Department of Health, Education and Welfare (HEW). 1931. Light Frame 

House Construction, Vocational Divisional Bulletin No. 145, U.S. 

Department of Health, Education and Welfare, Washington, DC.

U.S. Department of Housing and Urban Development (HUD). 1994. Alternative 

Framing Materials in Residential Construction: Three Case Studies.

Prepared by the NAHB Research Center, Inc., for the U.S. Department of 

Housing and Urban Development. Washington, DC: U.S. Department of 

Housing and Urban Development, Office of Policy Development and 

Research.
Residential Structural Design Guide 1-20

——. 1998a. Insulating Concrete Forms: Installed Cost and Acoustic 

Performance. Prepared by the NAHB Research Center, Inc., for the 

Portland Cement Association, National Association of Home Builders, and 

U.S. Department of Housing and Urban Development. Washington, DC: 

U.S. Department of Housing and Urban Development, Office of Policy 

Development and Research.

——. 1998b. Prescriptive Method for Insulating Concrete Forms in Residential 

Construction. Prepared by the NAHB Research Center, Inc., for the 

Portland Cement Association, National Association of Home Builders, and 

U.S. Department of Housing and Urban Development. Washington, DC:

U.S. Department of Housing and Urban Development, Office of Policy 

Development and Research.

Wagner, Willis H. 2003. Modern Carpentry—Building Construction Details in 

Easy-to-Understand Form, 10th ed. South Holland, IL: GoodheartWillcox Company, Inc.
Residential Structural Design Guide 2-1

CHAPTER 2

Structural Design 

Concepts

2.1 General

This chapter reviews some fundamental concepts of structural design and 

presents them in a manner relevant to the design of light-frame residential structures. 

Those concepts form the basis for understanding the design procedures and the 

overall design approach addressed in the remaining chapters of the guide. With this 

conceptual background, it is hoped that the designer will gain a greater appreciation 

for creative and efficient design of homes, particularly the many assumptions that 

must be made.

2.2 What Is Structural Design?

The process of structural design is simple in concept but complex in detail. It 

involves the analysis of a proposed structure to show that its resistance or strength 

will meet or exceed a reasonable expectation. That expectation usually is expressed 

by a specified load or demand and an acceptable margin of safety that constitutes a 

performance goal for a structure.

The performance goals of structural design are multifaceted. Foremost, a 

structure must perform its intended function safely over its useful life. Safety is 

discussed later in this chapter. The concept of useful life implies considerations of 

durability and establishes the basis for considering the cumulative exposure to timevarying risks (that is, corrosive environments, occupant loads, snow loads, wind 

loads, and seismic loads). Given that performance and cost are inextricably linked, 
Residential Structural Design Guide 2-2

however, owners, builders, and designers must consider economic limits to the 

primary goals of safety and durability.

Maintaining the appropriate balance between the two competing 

considerations of performance and cost is a discipline that guides the “art” of 

determining value in building design and construction. Value is judged by the “eye of 

the beholder,” however, and what is an acceptable value to one person may not be 

acceptable to another (for example, too costly versus not safe enough or not important 

versus important). For this reason, political processes factor into the development of 

minimum goals for building design and structural performance, with minimum value 

decisions embodied in building codes and engineering standards that are adopted as 

law. Design codes and standards applicable to engineered and prescriptive light-frame 

residential design are developed by an open consensus format. Changes are proposed, 

a public comment and discussion period is provided, and then a vote of eligible voters 

is taken.

In view of the preceding discussion, a structural designer seems to have little 

control over the fundamental goals of structural design, except to comply with or 

exceed the minimum limits established by law. Although this statement, in general, is 

true, a designer can still do much to optimize a design through alternative means and 

methods that call for more efficient analysis techniques, creative design detailing, and 

the use of innovative construction materials and methods. Structural designers have 

flexibility within a specific building code or design standard, depending on the exact 

wording. The National Design Specifications (NDS, 2010), for example, advise 

against designing a system in which a wood member is put into cross-grain bending, 

but NDS does not specifically prohibit that design if engineering and mechanics 

principles are applied. The FPL Wood Handbook (FPL, 2010) provides guidance 

about those types of situations. One such approach that has gained significant 

momentum, particularly for seismic design, is performance-based design (PBD). PBD 

allows designers to explicitly consider the performance of a building during design 

and usually focuses on extreme loadings, such as wind (van de Lindt and Dao, 2009) 

or earthquake events (FEMA, 2012; Filiatrault and Folz, 2002), but PBD has recently 

been proposed for other types of loading (van de Lindt et al., 2009).

Although the balance between cost and safety is, of course, paramount for 

many types of construction, including one- and two-family dwellings, structural 

designers can communicate to the owner (and other building stakeholders) that 

products and construction details are available that can improve building 

performance, and those options should be considered beyond the minimum design 

required by law (often referred to as “above code”). One such example would be to 

add hurricane clips (a metal connector sold by commercial suppliers) between the 

double top plate of the light-frame wall and the roof truss or joist, even though the 

clips may not be required by the building code. The added wind resistance would help 

ensure vertical load path continuity (discussed later in this guide) during strong, 

straightline winds and, potentially, small tornadoes (see, for example, Prevatt et al., 

2012 for further discussion).

In addition to exploring alternate means and methods such as PBD in the 

design of a residential wood-framed building, an engineered design calculated for a 

specific building configuration can be more cost effective than conventional 
Residential Structural Design Guide 2-3

construction design. Engineered design can be detailed to perform better and to 

address specific requirements, such as those for a building that will be constructed 

with heavy roofing materials or a site that has expansive soil conditions.

In summary, the goals of structural design are generally defined by law and 

reflect the collective interpretation of general public welfare by those parties involved 

in the development and local adoption of building codes. A designer’s role is to meet 

the goals of structural design as efficiently as possible and to satisfy a client’s 

objectives within the intent of the building code. The designer must bring to bear the 

fullest extent of his or her abilities, including creativity, knowledge, experience, 

judgment, ethics, and communication—aspects of design that are within the control 

of the individual designer and integral to a comprehensive approach to design. 

Structural design is much, much more than simply crunching numbers.

2.3 Load Types and Whole Building 

Response

The concepts presented in this section provide an overview of building loads 

and their effect on the structural response of typical wood-framed homes. As shown 

in table 2.1, building loads can be divided into two types, based on the orientation of 

the structural actions or forces that they induce: vertical loads and horizontal (that is, 

lateral) loads.

TABLE 2.1 Building Loads Categorized by Orientation

Vertical Loads Horizontal (Lateral) Loads

• Dead (gravity)

• Live (gravity)

• Snow (gravity)

• Wind (uplift on roof)

• Seismic and wind (overturning)

• Seismic (vertical ground motion)

• Wind

• Seismic (horizontal ground 

motion)

• Flood (static and dynamic 

hydraulic forces)

• Soil (active lateral pressure)

• Tsunami (dynamic hydraulic and 

forces)

2.3.1 Vertical Loads 

Gravity loads act in the same direction as gravity (that is, downward or 

vertically) and include dead, live, and snow loads. In general, they are static in nature 

and are usually considered a uniformly distributed or concentrated load. Tributary 

area is a term often used in design; it is the area of the building construction that is 

supported by a structural element, including the dead load (that is, weight of the 

construction) and the live load (that is, any applied loads). For example, the tributary 

gravity load on a floor joist would include the uniform floor load (dead and live) 

applied to the area of floor supported by the individual joist. The structural designer 
Residential Structural Design Guide 2-4

would select a standard beam or column model to analyze bearing connection forces 

(that is, reactions), internal stresses (that is, bending stresses, shear stresses, axial 

stresses, and deflection), and stability of the structural member or system (refer to 

appendix A for beam equations). The selection of an appropriate analytic model, 

however, is no trivial matter, especially if the structural system departs significantly 

from traditional engineering assumptions that are based on rigid body and elastic 

behavior. Such departures from traditional assumptions are particularly relevant to the 

structural systems that comprise many parts of a house, but to varying degrees.

Wind uplift forces are generated by negative (suction) pressures acting in an 

outward direction from the surface of the roof in response to the aerodynamics of 

wind flowing over and around the building. As with gravity loads, the influence of 

wind uplift pressures on a structure or assembly (that is, roof) is analyzed by using the 

concepts of tributary areas and uniformly distributed loads. The major differences 

between wind uplift and gravity loads are that wind pressures act perpendicular to the 

building surface (usually not in the direction of gravity) and that pressures can vary 

according to the size of the tributary area and its location on the building, particularly 

with proximity to changes in geometry (for example, eaves, corners, and ridges). 

Even though the wind loads are dynamic and highly variable, the design approach is 

based on a maximum static load (that is, pressure) equivalent.

Vertical forces also are created by overturning reactions that result from wind 

and seismic lateral loads acting on the overall building and its lateral force-resisting 

systems (LFRSs). Earthquakes also produce vertical ground motions or accelerations 

that increase the effect of gravity loads; however, vertical earthquake loads are 

usually implicitly addressed in the gravity load analysis of a light-frame building.

2.3.2 Lateral Loads 

The primary loads that produce lateral forces on buildings are attributable to 

forces associated with wind, earthquake ground motion, floods, soil, and, although 

rare, hurricane storm surge and tsunamis. Wind and earthquake lateral loads apply to 

the entire building. Lateral forces from wind are generated by positive wind pressures 

on the windward face of the building and by negative pressures on the leeward face of 

the building, creating a combined push-and-pull effect. Seismic lateral forces are 

generated by a structure’s dynamic inertial response to ground movement which 

reverses back and forth in an irregular cyclic motion. The magnitude of the seismic 

shear (that is, lateral) load depends on the intensity of the ground motion, the 

building’s mass, and the dynamic response characteristics of the building structure 

(that is, damping, ductility, stiffness, and so on). For houses and other similar lowrise 

structures, a simplified seismic load analysis employs equivalent static forces based 

on fundamental Newtonian mechanics (F = ma, or force = mass x acceleration), with 

adjustments to account for inelastic, ductile response characteristics of various 

building systems. Elevating structures on properly designed foundations can 

minimize flood loads, and avoiding building in a flood plain can eliminate flood loads 

altogether. Lateral loads from moving water and static hydraulic pressure are 

substantial. Soil lateral loads apply specifically to foundation wall design, mainly as 

an “out-of-plane” bending load on the wall.
Residential Structural Design Guide 2-5

Lateral loads also produce an overturning moment that must be offset by the 

dead load and connections of the building. Designers must, therefore, take into 

consideration the overturning forces on connections designed to restrain components 

from rotating or the building from overturning. Because wind is capable of generating 

simultaneous roof uplift and lateral loads, the uplift component of the wind load 

exacerbates the overturning tension forces that occur because of the lateral 

component of the wind load. Conversely, the dead load may be sufficient to offset the 

overturning and uplift forces, as is often the case in lower design wind conditions and 

in many seismic design conditions.

2.3.3 Structural Systems 

As far back as 1948, it was determined that “conventions in general use for 

wood, steel and concrete structures are not very helpful for designing houses because 

few are applicable” (NBS, 1948). More specifically, the National Bureau of Standards 

(NBS, now the National Institute of Standards and Technology) document 

encouraged the use of more advanced methods of structural analysis for homes. The 

International Residential Code (IRC; ICC, 2012) has made improvements over the 

past decade, providing some engineering-based prescriptive solutions for structural 

designers. These solutions, in turn, allow better consistency in reliability across 

different components and subassemblies. Most of the prescriptive provisions in the 

IRC, however, are based on conventional construction (this topic will be discussed in

more detail later in this chapter). Difficulties still exist in translating the results of 

studies of narrowly focused structural systems into general design applications for 

residential construction.

If a structural member is part of a system, as is typically the case in lightframe residential construction, its response is altered by the strength and stiffness 

characteristics of the system as a whole. In general, system performance includes two 

basic concepts known as load sharing and composite action. Load sharing is found in 

repetitive member systems (that is, wood framing) and reflects the ability of the load 

on one member to be shared by another or, in the case of a uniform load, the ability of 

some of the load on a weaker member to be carried by adjacent members. Composite 

action is found in assemblies of components that, when connected to one another, 

form a “composite member” with greater capacity and stiffness than the sum of the 

component parts.

The amount of composite action in a system depends on the manner in which 

the various system elements are connected. The aim is to achieve a higher effective 

section modulus than is provided by the individual component members. For 

example, when floor sheathing is nailed and glued to floor joists, the floor system 

realizes a greater degree of composite action than a floor with sheathing that is merely 

nailed; the adhesive between components helps prevent shear slippage, particularly if 

a rigid adhesive is used. Exact quantification of this result is difficult and beyond the 

scope of typical residential structural design. Slippage because of shear stresses 

transferred between the component parts necessitates consideration of partial 

composite action, which depends on the stiffness of an assembly’s connections.

Consideration of the floor as a system of fully composite T-beams, therefore, may 
Residential Structural Design Guide 2-6

lead to an unconservative solution, whereas the typical approach of considering only 

the floor joist member without taking into account the composite system effect will 

lead to a conservative design. For this reason, it is customary to consider the partial 

composite action of a glued-floor system only for computing deformation. Partial 

composite action is not considered for failure limit states.

This guide addresses the strength-enhancing effect of load sharing and partial 

composite action when information is available for practical design guidance. 

Repetitive-member increase factors (also called system factors) for general design 

have been quantified for a limited number of systems, such as floor systems and wall 

systems subjected to wind load. These system factors for general design use are 

necessarily conservative to cover a broad range of conditions. Exact quantification of 

system effects is a complex issue that would require extensive research, which has yet 

to be performed. 

System effects do not only affect the strength and stiffness of light-frame 

assemblies (including walls, floors, and roofs). They also alter the classical 

understanding of how loads are transferred among the various assemblies of a complex 

structural system, including a complete wood-framed home. For example, floor joists 

are sometimes doubled under non-load-bearing partition walls because of the added 

dead load and resulting stresses, determined in accordance with accepted engineering 

practice. Such practice is based on a conservative assumption regarding the load path 

and the structural response. That is, the partition wall creates an additional load but is 

relatively rigid and can also act as a deep beam, particularly when the top and bottom 

are attached to the ceiling and floor framing, respectively. As the floor is loaded and 

deflects, the interior wall helps resist the load. In engineered wood design, the 

reliability of the load path is reasonably known. Engineered wood design often makes 

simplifying assumptions to limit cost and effort for the designers. These simplifications 

are typically included because (1) without them, increased engineering cost could 

exceed construction savings, and (2) analysis tools are not readily available to model 

complicated three-dimensional structural configurations.

The preceding example of the composite action illustrates occasions where the 

interaction of separate structural systems or subassemblies results in improved 

structural response of the floor system such that it is able to carry more dead and live 

load than if the partition wall were absent. Whole-house assembly testing has 

demonstrated this effect (Hurst, 1965). Hence, a double joist should not be required 

under a typical non-load-bearing partition; in fact, a single joist may not even be 

required directly below the partition, assuming that the floor sheathing is adequately 

specified to support the partition between the joists. Although this condition cannot 

yet be duplicated in a standard analytic form conducive to simple engineering 

analysis, a designer should be aware of the concept when making design assumptions 

regarding light-frame residential construction.

Over the past 15 years, an increasing number of whole-house tests have been 

performed to better understand load transfer between components and subassemblies 

during system response. A comprehensive whole-structure test program was 

conducted in Australia by the Commonwealth Scientific and Industrial Research 

Organization (CSIRO) (Foliente et al., 2000a, 2000b; Paevere et al., 2000). Filiatrault 

et al. (2002) tested a two-story wood-framed house as part of the Consortium of 
Residential Structural Design Guide 2-7

Universities for Research in Earthquake Engineering-California Institute of 

Technology (CUREE-Caltech) project and later tested a larger two-story woodframed building (Christovasilis, Filiatrault, and Wanitkorkal, 2007; Filiatrault et al., 

2010) as part of the NEESWood project. Section 6.2 of the guide provides additional 

information on an array of whole-house tests conducted around the world.

At this point, consider that the response of a structural system, not just its 

individual elements, determines the manner in which a structure distributes and resists 

horizontal and vertical loads. For wood-framed systems, the departure from 

calculations based on classical engineering mechanics (that is, single members with 

standard tributary areas and assumed elastic behavior) and simplistic assumptions 

regarding load path can be substantial.

2.4 Load Path

Loads produce stresses on various systems, members, and connections as 

load-induced forces are transferred down through the structure to the ground. The 

path through which loads are transferred is known as the load path. A continuous load

path is capable of resisting and transferring the loads that are realized throughout the 

structure from the point of load origination down to the foundation.

As noted, the load path in a conventional home may be extremely complex 

because of the structural configuration and system effects that can result in substantial 

load sharing, partial composite action, and a redistribution of forces that depart from 

traditional engineering concepts. In fact, such complexity is an advantage that often 

goes overlooked in typical engineering analyses.

Further, because interior non-load-bearing partitions typically are neglected in 

a structural analysis, the actual load distribution will differ from that assumed in an 

elementary structural analysis. A strict accounting of structural effects would require 

numerical tools that are not widely available and are potentially too expensive, as 

mentioned previously. To the extent possible, a designer should consider system 

effects, recognizing that inherent uncertainties exist that may make the results 

imprecise.

2.4.1 Vertical Load Path 

Figures 2.1 and 2.2 illustrate vertically oriented loads created, respectively, by 

gravity and wind uplift. The wind uplift load originates on the roof from suction 

forces that act perpendicular to the exterior surface of the roof, as well as from 

internal pressure acting perpendicular to the interior surface of the roof-ceiling 

assembly in an outward direction. In addition, overturning forces resulting from 

lateral wind or seismic forces create vertical uplift loads (not shown in figure 2.2). In 

fact, a separate analysis of the lateral load path usually addresses overturning forces, 

necessitating separate overturning connections for buildings located in high-hazard 

wind or seismic areas (see section 2.3). As addressed in chapter 6, combining these 

vertical forces and designing a simple load path to accommodate wind uplift and 

overturning forces simultaneously may be feasible.
Residential Structural Design Guide 2-8

FIGURE 2.1 Vertical Load Path for Gravity Loads
Residential Structural Design Guide 2-9

FIGURE 2.2 Vertical Load Path for Wind Uplift
Residential Structural Design Guide 2-10

In a typical two-story home, the load path for gravity loads and wind uplift 

involves the following structural elements—

• Roof sheathing.

• Roof sheathing attachment.

• Roof framing member (rafter or truss).

• Roof-to-wall connection.

• Second-story wall components (top plate, studs, sole plate, headers, wall 

sheathing, and their interconnections).

• Second-story-wall-to-second-floor connection.

• Second-story-to-first-story-wall connection.

• First-story wall components (same as second story).

• First-story-wall-to-first-story or foundation connection.

• First-story-to-foundation connection.

• Foundation construction.

The preceding list makes obvious that numerous members, assemblies, and 

connections must be considered when tracking the gravity and wind uplift load paths in 

a typical wood-framed home. The load path itself is complex, even for elements such as 

headers that are generally considered simple beams. Usually, the header is part of a 

structural system (see figure 2.1), not an individual element single-handedly resisting 

the entire load originating from above. A framing system around a wall opening, not 

just a header, constitutes a load path.

Figure 2.1 also demonstrates the need for appropriately considering the 

combination of loads as the load moves “down” the load path. Elements that experience 

loads from multiple sources (for example, the roof and one or more floors) can be 

significantly overdesigned if design loads are not proportioned or reduced to account 

for the improbability that all loads will occur at the same time. Of course, the dead load 

is always present, but the live loads are transient; even when one floor load is at its 

lifetime maximum, the others will likely be at only a fraction of their design load.

Current design load standards generally allow for multiple transient load reductions; 

however, with multiple transient load-reduction factors intended for general use, those 

standards may not effectively address conditions relevant to a specific type of 

construction (that is, residential).

Consider the soil-bearing reaction at the bottom of the footing in figure 2.1. As 

implied by the illustration, the soil-bearing force is equivalent to the sum of all tributary 

loads, dead and live. However, it is important to understand the combined load in the 

context of design loads. Floor design live loads are based on a lifetime maximum 

estimate for a single floor in a single level of a building, but the occupancy conditions on 

the upper and lower stories in homes typically differ. When one load is at its maximum, 

the other is likely to be at a fraction of its maximum. Designers are able to consider the 

live loads of the two floors as separate transient loads; specific guidance is available in 

ASCE 7–10 (ASCE, 2010). In concept, the combined live load should be reduced by an 

appropriate factor, or one of the loads should be set at a point-in-time value that is a 

fraction of its design live load. For residential construction, the floor design live load is 

either 30 pounds per square foot (psf; for bedroom areas) or 40 psf (for other areas), 
Residential Structural Design Guide 2-11

although some codes require a design floor live load of 40 psf for all areas. In contrast, 

average sustained live loads during typical use conditions are about 6 psf (with one 

standard deviation of 3 psf), which is about 15 to 20 percent of the design live load 

(Chalk and Corotis, 1980). If actual loading conditions are not rationally considered in a 

design, the result may be excessive footing widths, header sizes, and so forth.

When tracking the wind uplift load path (figure 2.2), the designer must consider 

the offsetting effect of the dead load as it increases down the load path. Building codes 

and design standards, however, do not permit the consideration of any part of the 

sustained live load in offsetting wind uplift, even though some minimum point-in-time 

value of floor live load is likely present if the building is in use—that is, furnished or 

occupied. In addition, other “nonengineered” load paths, such as those provided by 

interior walls and partitions, are not typically considered. Although these are prudent 

limits, they help explain why certain structures may not “calculate” but otherwise 

perform adequately.

Building codes commonly consider only 0.6 of the dead load when analyzing a 

structure’s net wind uplift forces. The 0.6 factor is a way of preventing the potential 

error of requiring insufficient connections where a zero uplift value is calculated in 

accordance with a nominal design wind load (as opposed to the ultimate wind event 

that is implied by the use of a safety margin for material strength in unison with a 

nominal design wind speed). Furthermore, building code developers have expressed a 

concern that engineers might overestimate actual dead loads, which would be 

conservative when designing members for gravity loads but un conservative when 

designing members for combined dead and wind loads.

For complicated house configurations, a load of any type may vary considerably 

at different points in the structure, necessitating a decision of whether to design for the 

worst case or to accommodate the variations. Often the worst case condition is applied to 

the entire structure even when only a limited part of the structure is affected. For 

example, a floor joist or header may be sized for the worst case span and used throughout 

the structure. The worst case decision is justified only when the benefit of a more 

intensive design effort is not offset by a significant cost reduction. Another important 

consideration is the more detailed analysis of various design conditions that usually 

results from greater construction complexity. Simplification and cost reduction are both 

important design objectives, but they may often be mutually exclusive. The consideration 

of system effects in design, as discussed previously, may result in both simplification and 

cost efficiencies that improve the quality and affordability of the finished product.

One helpful attribute of traditional platform-framed home construction is that 

the floor and roof gravity loads are typically transferred through bearing points, not 

connections. Thus, connections may contribute little to the structural performance of 

homes with respect to vertical loads associated with gravity (that is, dead, live, and 

snow loads).

By contrast, metal plate-connected roof and floor trusses rely on connections to 

resist gravity loads, but these engineered components are designed and produced in 

accordance with a proven standard and are generally highly reliable (TPI, 2007). 

Indeed, the metal plate-connected wood truss was first conceived in Florida in the 

1950s to respond to the need for improved roof structural performance, particularly 

with respect to connections in roof construction (Callahan, 2002).
Residential Structural Design Guide 2-12

In high-wind climates, where the design wind uplift load approaches offsetting 

the actual dead load, the consideration of connection design in wood-framed assemblies 

becomes critical for roofs, walls, and floors (the dead load used to offset wind uplift is 

the actual dead load, not the design dead load). In fact, the importance of connections in 

conventionally built homes is evidenced by the common loss of weakly attached roof 

sheathing or roofs in extreme wind events, such as moderate- to large-magnitude 

hurricanes.

Newer prescriptive code provisions have addressed many of the historic structural 

wind damage problems by specifying more stringent general requirements (AWC, 2012; 

ICC, 2012). In many cases, the newer high-wind prescriptive construction requirements 

may be improved by more efficient site-specific design solutions that consider wind 

exposure and system effects and that include other analytic improvements. Site-specific 

design solutions may also improve prescriptive seismic provisions in the latest building 

codes for conventional residential construction (ICC, 2012).

2.4.2 Lateral Load Path

The overall system that provides lateral resistance and stability to a building is 

known as the LFRS. In light-frame construction, the LFRS includes shear walls and 

horizontal diaphragms. Shear walls are walls that are typically braced or clad with 

structural sheathing panels to resist racking forces. Horizontal diaphragms are floor 

and roof assemblies that are also usually clad with structural sheathing panels. 

Although more complicated and difficult to visualize, the lateral forces imposed on a 

building from wind or seismic action also follow a load path that distributes and 

transfers shear and overturning forces from lateral loads. The lateral loads of primary 

interest are those resulting from—

• The horizontal component of wind pressures on the building’s exterior 

surface area.

• The inertial response of a building’s mass and structural system to 

earthquake ground motions. 

As seen in figure 2.3, the lateral load path in wood-framed construction 

involves entire structural assemblies (that is, walls, floors, and roofs) and their 

interconnections, not just individual elements or frames, as would be the case with 

typical steel or concrete buildings that use discrete braced framing systems. The 

distribution of loads in figure 2.3’s three-dimensional load path depends on the 

relative stiffness of the various components, connections, and assemblies that 

constitute the LFRS. To complicate the problem further, stiffness is difficult to 

determine because of the nonlinearity of the load-displacement characteristics of 

wood-framed assemblies and their interconnections. Figure 2.4 illustrates a deformed 

light-frame building under lateral load (the deformations are exaggerated for 

conceptual purposes). Note, however, that American Society of Civil Engineers 

(ASCE) 7 (ASCE, 2010) does not require that torsion be included for wind load 

analyses for light-frame construction that is two stories or less; it is required only in 

seismic analyses.
Residential Structural Design Guide 2-13

FIGURE 2.3 Lateral Load Path
Residential Structural Design Guide 2-14

FIGURE 2.4 Building Deformation Under Lateral Load
Residential Structural Design Guide 2-15

Lateral forces from wind and seismic loads also create overturning forces that 

cause a “tipping” or “rollover” effect. When these forces are resisted, a building is 

prevented from overturning in the direction of the lateral load. On a smaller scale, 

overturning forces are also realized at the shear walls of the LFRS such that the shear 

walls must be restrained from rotating or rocking on their base by proper connection. 

This is often done with anchor bolts or hold down hardware. On an even smaller 

scale, the forces are realized in the individual shear wall segments between openings 

in the walls. 

The overturning force diagrams in Figure 2.3 are based on conventionally 

built homes constructed without hold-down devices positioned to restrain shear wall 

segments independently. It should be noted that the effect of dead loads that may 

offset the overturning force and of wind uplift loads that may increase the overturning 

force is not necessarily depicted in Figure 2.3’s conceptual plots of overturning forces 

at the base of the walls. If rigid steel hold-down devices are used in designing the 

LFRS, the wall begins to behave in a manner similar to a rigid body at the level of 

individual shear wall segments, particularly when the wall is broken into discrete 

segments as a result of the configuration of openings in a wall line.

Significant judgment and uncertainty attend the design process for 

determining building loads and resistance, including definition of the load path and 

the selection of suitable analytic methods. This guide is intended to serve as a 

resource for designers who are considering the use of alternative analytic methods 

when current approaches may not adequately address the design issue.

2.5 Structural Reliability

Before addressing the “nuts and bolts” of the structural design of singlefamily dwellings, one must understand the fundamental concept of structural 

reliability. Although safety is generally based on the rational principles of risk and 

probability theory known as structural reliability, it is also subject to some level of 

judgment, particularly the experience and understanding of those who participate in 

the development of building codes and design standards. Slight differences exist in 

the various code-approved sources for design loads, load combinations, load factors, 

and other features that can affect structural safety. National load and material design 

standards, however, have established a consistent basis for safety in structural design. 

It should be noted that residential occupancies are considered in the establishment of 

loads. Most importantly, the aim of any design approach is to ensure that the 

probability of failure (that is, load exceeding resistance) is acceptably small or, 

conversely, that the level of reliability is sufficiently high.

A common misconception is that design loads alone determine the amount of 

“safety” achieved in a design. For example, a typical conclusion reached in the 

aftermath of Hurricane Andrew was that the storm’s wind speed exceeded the design 

wind speed map value; therefore, the wind map (used as the source for the design 

load) was perceived to be insufficient. In other cases, such as the Northridge 

Earthquake, reaction to various anecdotal observations resulted in increased safety 

factors for certain materials (that is, wood design values were decreased by 25 percent 

by the City of Los Angeles). In reality, numerous factors affect the level of reliability 
Residential Structural Design Guide 2-16

in a structural system, just as several factors determine the level of performance 

realized by buildings in a single extreme event, such as Hurricane Andrew or the 

Northridge Earthquake.

Structural reliability is a multifaceted performance goal that integrates all 

objective and subjective aspects of the design process, including the following major 

variables—

• Determination of characteristic material or assembly strength values based 

on tested material properties and their variabilities.

• Application of a nominal or design load based on a statistical 

representation of load data and the data’s uncertainty or variability.

• Consideration of various uncertainties associated with the design practice 

(for example, competency of designers and accuracy of analytic 

approaches), the construction practice (for example, quality or 

workmanship), and durability.

• Selection of a level of reliability that considers the preceding factors and 

the consequences of exceeding a specified design limit state (that is, 

collapse, deformation, or the onset of “unacceptable” damage).

When the aforementioned variables are known or logically perceived, many 

ways are available to achieve a specified level of safety. As a practical necessity, 

however, the design process has been standardized to provide a reasonably consistent 

basis for applying the following key elements of the design process—

• Characterizing strength properties for various material types (for example, 

steel, wood, concrete, and masonry).

• Defining nominal design loads and load combinations for crucial inputs 

into the design process.

• Conveying an acceptable level of safety (that is, a safety margin) that can 

be easily and consistently applied by designers. 

Institutionalized design procedures provide a basis for selecting from the vast 

array of structural material options available in the construction market. The 

generalizations necessary to address the multitude of design conditions, however, rely 

on a simplified and standardized format and thus often overlook special aspects of a 

particular design application.

The following sections discuss safety, but they are intentionally basic and 

focus on providing the reader with a conceptual understanding of safety and 

probability as a fundamental aspect of engineering. Probability concepts are 

fundamental to modern design formats, such as load and resistance factor design 

(LRFD), which is also known as reliability-based design or strength design. The same 

concepts are also crucial to understanding the implications of the simple safety factor 

in traditional allowable stress design (ASD). In 2002, the Committee on ReliabilityBased Design of Wood Structures undertook a special project for the ASCE 

Structural Engineering Institute (SEI). The objective was to quantify the reliability 

inherent in AF&PA/ASCE 16 (1996) using state-of-the-art structural reliability 
Residential Structural Design Guide 2-17

methods. The project resulted in a series of papers (Bulleit et al., 2004; Rosowsky et 

al., 2004; van de Lindt and Rosowsky, 2005). Several years later, the same committee 

completed another SEI special project that examined the feasibility of applying PBD 

principles to wood design (see chapter 1) (van de Lindt et al., 2009).

That study addressed both the benefits and the challenges. As discussed 

previously in this chapter, PBD concepts will be mentioned throughout this guide as 

an option for the structural designer to improve the performance of residential 

structures. Following are some additional references.

• Probability Concepts in Engineering Planning and Design. Vol. I, Basic 

Principles (Ang and Tang, 1975).

• CRC Structural Engineering Handbook. chap. 29, “Structural Reliability” 

(Chen, 1997).

• Probabilistic Structural Mechanics Handbook: Theory and Industrial 

Applications (Sundararajan, 1995).

• Uncertainty Analysis, Loads, and Safety in Structural Engineering (Hart, 

1982).

• Statistical Models in Engineering (Hahn and Shapiro, 1967).

• Reliability of Structures, 2nd Ed. (Nowak and Collins, 2013).

2.5.1 Nominal Design Loads

Nominal design loads are generally specified on the basis of probability, with 

the interchangeable terms “return period” and “mean recurrence interval” often used 

to describe the probability of loads. Either term represents a condition that is 

predicted to be met or exceeded once, on average, during the reference time period. 

For design purposes, loads are generally evaluated in terms of annual extremes (that 

is, the variability of the largest load experienced in any given 1-year period) or 

maximum lifetime values.

The historical use of safety factors in ASD has generally been based on a 50-

year return period design load. With the advent of LRFD, the calculation of nominal 

loads has shifted away from ASD for some load types. Now, earthquake and wind 

design use design values represented by hazard levels considered to be ultimate (or 

LRFD level) events. The Maximum Considered Earthquake is the intensity of ground 

motion that has the probability of exceedance of 2 percent in 50 years (for example, a 

2,500-year return period). Earthquake design loads are based on a 2/3 factor of the 

ground motion that occurs during the 2,500-year event. They are computed from 

annual probabilities and design periods and is expressed as P = 1- (1-Pa)

n where Pa is 

the annual probability (1/return period), P is the probability of exceedance during the 

time period of interest, and n is the time period of interest. This formula is described 

in the commentary of ASCE 7–10.

ASCE 7–10 (ASCE, 2010) provides risk-targeted seismic design maps for the 

conterminous United States (Luco et al., 2007). One key result of the move from 

uniform-hazard to risk-targeted mapped spectral accelerations is a reduction in the 

design spectral acceleration for the central and eastern United States to 70 to 90 
Residential Structural Design Guide 2-18

percent of their 2005 values. This reduction occurred because previous mapping 

considered only the magnitude of the event, not the likely frequency.

The method of determining a design load also differs according to the type of 

load and the availability of data to evaluate the time-varying nature of loads. The 

derivation of various nominal loads may be assembled from information and 

references contained in the ASCE 7 standard (ASCE, 2010). Design wind loads are 

based on a probabilistic analysis of wind speed data collected from many weather 

stations across the United States. The data include wind loads in most of the country 

and hurricane simulation modeling for wind speeds along the hurricane-prone 

coastlines. The wind speed maps in ASCE 7–10 represent the speeds that have a 7-

percent probability of exceedance in 50 years, or a 700-year return period for 

residential structures (see section 3.6 on wind design). Snow loads are based on 

snowfall or ground snow depth data and are correlated to roof snow loads through 

recent studies. Snow drift loads in ASCE 7–10 (ASCE, 2010) have improved from 

earlier versions of the standard by adding a new thermal factor and by not requiring 

unbalanced snow loads be applied to hip and gable roofs when the roof slope is 

steeper than 7 on 12 or is shallower than 1⁄2 on 12 (1/2:12). 

Earthquake loads are defined from historical ground motion data and 

conceptualized risk models based on direct or indirect evidence of past earthquake 

activity. The maps that illustrate the seismic ground motion have been developed by 

the U.S. Geological Survey. Considerable uncertainty exists in the estimation of 

seismic hazards, particularly in areas that are believed to have low seismicity (that is, 

few events) but the potential for major seismic events. Details of the ASCE 7–10 map 

development can be found in Luco et al. (2007). Floor live loads are modeled by 

using live load surveys of “point-in-time” loading conditions and hypotheses or 

judgment concerning extreme or maximum lifetime loads. In some cases, expert 

panels decide on appropriate loads or related load characteristics when adequate data 

are not available.

In summary, the determination of load characteristics is based on historical 

data, risk modeling, and expert opinion. Those factors, in turn, guide the specification 

of nominal design loads for general design purposes in both the ASD and LRFD 

formats. It is important to remember that the return period of the design load is not 

the only factor determining safety; the selection of safety factors (ASD), load factors 

(LRFD), or performance objectives depends on the definition of a nominal design 

load (that is, its return period) and the material’s strength characterization to achieve a 

specified level of safety.

2.5.2 Basic Safety Concepts in Allowable

Stress Design

The concept of ASD is demonstrated in a generic design equation or 

performance function (see equation 2.5-1) for a wood framing member. A common 

practice in traditional ASD is to divide the characteristic (for example, fifth 

percentile) material strength value by a safety factor of greater than 1 to determine an 

allowable design strength that is dependent on a selected limit state (that is, a 

proportional limit or rupture) and material type, among other factors that involve the 
Residential Structural Design Guide 2-19

judgment of specification-writing groups. Most factors of safety fall in the range of 

1.5 to 2.5 for residential design. The allowable design strength is then compared to 

the stresses created by a nominal design load combination, usually based on a 50-year 

mean recurrence interval. A lower safety factor is generally applied to design 

conditions that are less variable or that are associated with a “noncritical” 

consequence, while the higher safety factor is typically applied to elements associated 

with greater uncertainty, such as connections. In addition, a higher safety factor is 

usually selected for materials, systems, or stress conditions that result in an abrupt 

failure mode without warning. The safety factor is also intended to cover the 

variability in loads in ASD.

Equation 2.5-1

where,

R = the nominal resistance (or design stress), usually based 

on the fifth percentile strength property of interest (also 

known as the characteristic strength value).

S.F. = the safety factor (R/S.F. is known as the allowable 

stress).

L = the load effect caused by the nominal design load 

combination (in units of R).

The equation refers to characteristic material strength, which represents the 

material stress value used for design purposes (also known as nominal or design 

strength or stress). When characteristic material strength (normalized to standard 

conditions) is divided by a safety factor, the result is an allowable material strength or 

stress. Given that materials exhibit variability in their stress capacity (some more 

variable than others), it is necessary to select a statistical value from the available 

material test data. Generally, but not always, the test methods, data, and evaluations 

of characteristic material strength values follow standardized procedures that vary 

across material industries (for example, concrete, wood, and steel) in part because of 

the uniqueness of each material. In most cases, the characteristic strength value is 

based on a lower bound test statistic such as the fifth percentile, which is a value at 

which no more than 5 percent of the material specimens from a sample exhibit a 

lesser value. Because sampling is involved, the sampling methodology and sample 

size become critical to confidence in the characteristic strength value for general 

design applications.

In some cases, procedures for establishing characteristic material strength 

values are highly sophisticated and address many of the concerns mentioned 

previously; in other cases, the process is simple and involves reduced levels of 

exactness or confidence (for example, use of the lowest value in a small number of 

tests). Generally, the more variable a material, the more sophisticated the 

determination of characteristic material strength properties. A good example is the 

wood industry, whose many species and grades of lumber further complicate the 

L

S.F.

R

≥
Residential Structural Design Guide 2-20

inherent nonhomogeneity of the product. The wood industry, therefore, uses fairly 

sophisticated procedures to sample and determine strength properties for a multitude 

of material conditions and properties (see chapter 5).

2.5.3 Basic Safety Concepts in Load and Resistance 

Factor Design

The LRFD format has been conservatively calibrated to the level of safety 

represented by past ASD design practice and thus retains a tangible connection 

with historically accepted norms of structural safety (Ellingwood et al., 1982; 

Galambos et al., 1982; and others);1 thus, either method achieves a similar level of 

safety. The LRFD approach, however, uses two factors—one applied to the load 

and one applied to the resistance or strength property—that permits more 

consistent treatment of safety across a broader range of design conditions.

Equation 2.5-2 shows, conceptually, the LRFD design format (that is, 

performance function) and compares a factored characteristic resistance value 

with a factored nominal load. Thus, for a given hazard condition and given 

material—and similar to the outcome described in the previous section on ASD—

increasing the load factor or decreasing the resistance factor has the effect of 

increasing the level of safety. Figure 2.5 depicts the variable nature of building 

loads and resistance and the safety margin relative to design loads and nominal 

resistance.

 1

Historically accepted performance of wood-framed design, particularly housing, has not been specially considered 

in the development of modern LRFD design provisions for wood or other materials (such as concrete in 

foundations).
Residential Structural Design Guide 2-21

Equation 2.5-2 φ

where,

φ = resistance factor (phi).

R = nominal resistance or design stress, usually based on the 

fifth percentile strength property of interest (also known as 

the characteristic strength value).

γ = load factor for each load in a given load combination 

(gamma).

L = the stress created by each load in a nominal design load 

combination (in units of R).

A resistance factor is applied to a characteristic material strength value to 

account for variability in material strength properties. The resistance factor generally 

ranges from 0.5 to 0.9, with the lower values applicable to those strength properties 

that have greater variability or that are associated with an abrupt failure (one with 

little warning). The resistance factor also depends on the selected characterization of 

the nominal or characteristic strength value for design purposes (that is, average, 

lower fifth percentile, lowest value of a limited number of tests, and so on).

A load factor is individually applied to each load in a nominal design load 

combination to account for the variability and nature of the hazard or combined 

hazards. It also depends on the selected characterization of the nominal load for 

design purposes (for example, 50-year return period, 475-year return period, or 

others). In addition, the load factors proportion the loads relative to each other in a 

combination of loads (that is, account for independence or correlation between loads 

and their likely “point-in-time” values when one load assumes a maximum value). 

Thus, the load factor for a primary load in a load combination is generally 1.0 in 

LRFD. For other transient loads in a combination, the factors are generally much less 

than 1. In this manner, the level of safety for a given material and nominal design 

load is determined by the net effect of factors—one on the resistance side of the 

design equation and the others on the load side. For ASD, the factors and their 

purpose are embodied in one simple element the safety factor.

R ≥ ∑ γL
Residential Structural Design Guide 2-22

FIGURE 2.5 Basic Concept of Safety in LRFD and ASD Considering the 

Variability of Loads and Resistance

2.5.4 Basic Safety Concepts in Performance-Based Design

PBD is a design approach or methods that allow the designers (or team) to 

explicitly consider performance objectives during the design process. An ASCE 

special project, titled “The Next Step for AF&PA/ASCE 16: Performance-Based 

Design of Wood Structures,” was recently completed (van de Lindt et al., 2009). PBD 

has been documented for several decades and has its origin in fire engineering, for 

which the objective is product development that meets a particular prescribed 
Residential Structural Design Guide 2-23

performance, for example, a 1-hour fire rating. Earthquake engineering followed suit 

with the Structural Engineers Association of California Vision 2000 (1996) document 

after the 1989 Loma Prieta and 1994 Northridge earthquakes. The primary seismic 

PBD methodology is described in FEMA P-58, Seismic Performance Assessment of 

Buildings: Methodology and Implementation (FEMA, 2012). The difference in this 

design approach is in the consideration of outcomes of the design when a defined 

hazard level determines the design of the building. Designs based on building codes 

follow defined criteria in the codes that are intended to provide some level of 

performance; however, that performance level is never stated and is seldom evaluated 

except when the building is affected by a design event (FEMA, 2012). Designs based 

on performance are based on desired outcomes and levels of building performance 

during and after an event occurs. The beginning points for PBD are the needs of the 

building owners or stakeholders, not the requirements of the building code.

Although not in widespread use as of this revision of the guide, PBD for wind 

engineering has been envisioned (van de Lindt and Dao, 2009). Work has begun in 

wind engineering using as a starting point the process developed for the seismic 

hazard. The building designs that follow will likely be different for the two hazards, 

given that seismic designs usually are driven by collapse prevention techniques for 

extreme events, and wind designs may be driven by preventing weather penetration 

into the building envelope.

2.5.5 Putting Safety and Performance into Perspective

Safety is a relative measure that must be interpreted in consideration of the 

many assumptions underlying the treatment of uncertainty in the design process. Any 

reliable measure of safety must look to past experience and attempt to evaluate 

historic data in a rational manner to predict the future. Economic consequences are 

becoming increasingly debated and influential in the development of codified 

guidelines for structural design, which, as discussed previously, has led to the 

development of PBD in the seismic arena. Of course, such a design philosophy 

explicitly considers the performance objectives for a structure.

Implicit consideration of building performance also has been routinely 

achieved through improved building codes and standards. For example, following 

Hurricane Charley (2004), the Institute for Business and Home Safety (IBHS) 

concluded that “enforcement of modern engineering design based building codes 

made a positive impact on the performance of residential homes during Hurricane 

Charley in 2004. The frequency of claims was reduced by 60 percent and the claim 

was 42 percent less severe when a loss did occur, for homes built after the adoption of 

the modern codes,” (IBHS, 2004: pg. 5). For more information on residential building 

codes and enforcement, see IBHS (2011). Thus, some engineering requirements in 

codes may address two very different objectives—life safety and property protection 

or damage reduction. Finally, the manner in which these two different forms of risk 

are presented can have a profound impact on the perspective of risk and the perceived 

need for action or inaction.
Residential Structural Design Guide 2-24

2.6 References

American Wood Council (AWC). 2012. Wood Frame Construction Manual. 

Washington, DC: American Wood Council.

———. 2010. National Design Specifications (NDS). Washington, DC: American 

Wood Council.

Ang, Alfredo H. S., and Wilson H. Tang. 1975. Probability Concepts in Engineering 

Planning and Design. Vol. I, Basic Principles. New York: John Wiley & 

Sons.

American Society of Civil Engineers (ASCE). 2010. Minimum Building Loads for 

Buildings and Other Structures (ASCE 7–10). Reston, VA: American Society 

of Civil Engineers.

Bulleit et al. 2005. “Modeling Wood Walls Subjected to Combined Transverse and 

Axial Loads.” Journal of Structural Engineering 131, No. 5: 781–793.

Reston, VA: ASCE.

Callahan, Edward E. 2002. Metal Plate Connected Wood Truss Handbook. 3rd ed. 

Madison, WI: Wood Truss Council of America.

Chalk, P. and R. Corotis. 1980. “Probability Model for Design Live Loads.” Journal 

of the Structural Division 106, No. 10: 2017-2033. Reston, VA: ASCE. 

Chen, W.F. 1997. “Structural Reliability.” Chap. 29 in Structural Engineering 

Handbook. Boca Raton, FL: CRC Press.

Christovasilis, I. P., A. Filiatrault, and A. Wanitkorkal. 2007. “Seismic Test of a FullScale Two-Story Light-Frame Wood Building: The NEESWood Benchmark 

Test.” NEESWood report NW-01, MCEER, Buffalo, NY.

Ellingwood, B., J. G. MacGregor, T. V. Galambos, and C. A. Cornell. 1982. 

“Probability Based Load Criteria: Load Factors and Load Combinations.” 

Journal of the Structural Division, 108, No. 5: 978–997.

Federal Emergency Management Agency (FEMA). 2012. Seismic Performance 

Assessment of Buildings: Methodology and Implementation (FEMA P-58). 

Washington, DC: FEMA.

Filiatrault, A., and Folz, B., 2002. “Performance-Based Seismic Design of Wood 

Framed Buildings”, Journal of Structural Engineering, 128, No.1, 39-47.
Residential Structural Design Guide 2-25

Filiatrault, A., D. Fischer, B. Folz, and C-M. Uang. 2002. “Seismic Testing of a TwoStory Woodframe House: Influence of Wall Finish Materials.” Journal of 

Structural Engineering 128, No. 10: 1337–1345.

Filiatrault, A., I. P. Christovasilis, A. Wanitkorkal, and J. W. van de Lindt. 2010. 

“Experimental Seismic Response of a Full-Scale Light-Frame Wood 

Building.” Journal of Structural Engineering 136, No. 3: 246–254.

Foliente, G. C., B. Kasal, P.J. Paevere, and M. Collins. 2000a. “Whole Structure 

Testing and Analysis of a Light-Frame Wood Building: Phase II—Design 

Procedures against Lateral Loads. BCE Doc 00/177. CSIRO, Highett, 

Victoria, Australia.

Foliente, G. C., B. Kasal, P. J. Paevere, L. Macindoe, R. Banks, M. Syme, C. Seath, 

and M. Collins. 2000b. “Whole Structure Testing and Analysis of a LightFrame Wood Building: Phase I—Test House Details and Preliminary 

Results.” BCE Doc 00/033. CSIRO, Highett, Victoria, Australia.

Forest Products Laboratory. 2010. “Wood handbook—Wood as an engineering

material.” General Technical Report FPL-GTR-190. Madison, WI: U.S.

Department of Agriculture, Forest Service, Forest Products Laboratory.

Galambos, T. V., et al.. 1982. “Probability Based Load Criteria: Assessment of 

Current Design Practice.” Journal of the Structural Division 108, No. 5: 959–

977.

Hahn, G. J., and S. S. Shapiro. 1967. Statistical Models in Engineering. New York: 

John Wiley & Sons.

Hart, Gary C. 1982. Uncertainty Analysis, Loads, and Safety in Structural 

Engineering. Englewood Cliffs, NJ: Prentice-Hall, Inc.

Hurst, Homer T. 1965. The Wood-Frame House as a Structural Unit. Technical 

Report No. 5, National Forest Products Association, Washington, DC.

Institute for Business and Home Safety (IBHS). 2004. “Hurricane Charley: Nature’s 

Force vs. Structural Strength—Executive Summary” Hurricane Charley PostHurricane Report, IBHS, Tampa, FL.

———. 2011. “Rating the States: An Assessment of Residential Building Code and 

Enforcement Systems for Life Safety and Property Protection in HurricaneProne Regions.” IBHS.

International Code Council (ICC). 2012. International Residential Code. Washington, 

DC: ICC.
Residential Structural Design Guide 2-26

Luco, N., B. R. Ellingwood, R. O. Hamburger, J. D. Hooper, and J. K. Kimball. 2007. 

Proceedings of the 2007 Convention of the Structural Engineers Association 

of California, Sacramento, CA. 

National Bureau of Standards (NBS). 1948. Strength of Houses: Application of 

Engineering Principles to Structural Design. Building Materials and 

Structures Report 109, U.S. Department of Commerce, National Bureau of 

Standards. Washington, DC: GPO.

Nowak, A. S., and K. R. Collins. 2013. Reliability of Structures, 2nd ed. Boca Raton, 

FL: CRC Press.

Paevere, P. J., G. C. Foliente, L. Macindoe, R. Banks, M. Syme, C. Seath, B. Kasal, 

and M. Collins. 2000. “Whole Structure Testing and Analysis of a LightFrame Wood Building: Phase III—Destructive Testing.” BCE Doc 00/425. 

CSIRO, Highhett, Victoria, Australia.

Prevatt et al. 2012. Making the case for improved structural design: Tornado 

outbreaks of 2011. Leadership and Management in Engineering 12 (4), 254-

270.

Rosowsky, D. V., Y. Guangren, and W. M. Bulleit, 2005. “Reliability of Light-Frame 

Wall Systems Subject to Combined Axial and Transverse Loads.” Journal of 

Structural Engineering, 131, No. 9: 1444-1455.

Structural Engineers Association of California Vision 2000. 1995. SEAOC (1995). 

"Vision 2000—A Framework for Performance Based Earthquake 

Engineering." Vol. 1, January 1995.

Sundararajan, C. 1995. Probabilistic Structural Mechanics Handbook: Theory and 

Industrial Applications. New York: Chapman & Hall.

Truss Plate Institute (TPI). 2007. TPI 1: National Design Standard for Metal Plate 

Connected Wood Truss Construction. Truss Plate Institute, Madison, WI.

van de Lindt, J. W., Y. Li, W.M. Bulleit, R. Gupta, and P. I. Morris. 2009. “The Next 

Step for ASCE 16: Performance-Based Design of Wood Structures.” Journal 

of Structural Engineering, 135, No. 6: 611–618.

van de Lindt, J.W., Y. Li, W.M. Bulleit, R. Gupta, and P. I. Morris. (2009). “The 

Next Step for ASCE 16: Performance-Based Design of Wood Structures.” 

Journal of Structural Engineering, 135, No. 6: 611- 618.

van de Lindt, J. W. and T. N. Dao. 2009. “Performance-Based Wind Engineering for 

Wood Frame Buildings.” Journal of Structural Engineering, 135, No. 2: 169–

177.
Residential Structural Design Guide 2-27

van de Lindt, J.W. and D. V. Rosowsky. 2005. “Strength-Based Reliability of Wood 

Shearwalls Subjected to Wind Load.” Journal of Structural Engineering, 131, 

No. 2: 359-363.
Residential Structural Design Guide 3-1

CHAPTER 3

Design Loads for 

Residential Buildings

3.1 General

Loads are a primary consideration in any building design because they define the nature and 

magnitude of hazards or external forces that a building must resist to provide reasonable 

performance (that is, safety and serviceability) throughout the structure’s useful life. The 

anticipated loads are influenced by a building’s intended use (occupancy and function), 

configuration (size and shape), and location (climate and site conditions). Ultimately, the type 

and magnitude of design loads affect critical decisions such as material selection, construction 

details, and architectural configuration. To optimize the value (that is, performance versus 

economy) of the finished product, therefore, design loads must be applied realistically.

Although the buildings considered in this guide are primarily single-family detached and 

attached dwellings, the principles and concepts related to building loads also apply to other 

similar types of construction, such as low-rise apartment buildings. In general, the design loads 

recommended in this guide are based on applicable provisions of the ASCE 7 standard–

Minimum Design Loads for Buildings and Other Structures (ASCE, 2010). The ASCE 7 

standard represents an acceptable practice for building loads in the United States and is 

recognized in U.S. building codes. For this reason, the reader is encouraged to become familiar 

with the provisions, commentary, and technical references contained in the ASCE 7 standard.

In general, the structural design of housing has not been treated as a unique engineering 

discipline or subjected to a special effort to develop better, more efficient design practices. For 

that reason, this part of the guide focuses on those aspects of ASCE 7 and other technical 

resources that are particularly relevant to the determination of design loads for residential 

structures. The guide provides supplemental design assistance to address aspects of residential 

construction for which current practice is either silent or in need of improvement. The guide’s 
Residential Structural Design Guide 3-2

methods for determining design loads are complete yet tailored to typical residential conditions. 

As with any design function, the designer must ultimately understand and approve the loads for 

a given project as well as the overall design methodology, including all its inherent strengths 

and weaknesses. Because building codes from different jurisdictions can vary in their treatment 

of design loads, the designer should, as a matter of due diligence, identify variances from both 

local accepted practice and the applicable building code relative to design loads as presented in 

this guide, even though the variances may be considered technically sound.

Complete design of a home typically requires the evaluation of several different types of 

materials, as discussed in chapters 4 through 7. Some material specifications use the allowable 

stress design (ASD) approach while others use load and resistance factor design (LRFD). 

Chapter 4 uses the LRFD method for concrete design and the ASD method for masonry 

design. For wood design, chapters 5, 6, and 7 use ASD. For a single project, therefore, the 

designer may have to determine loads in accordance with both design formats. This chapter 

provides load combinations intended for each method. The determination of individual 

nominal loads is essentially unaffected. Special loads, such as ice loads and rain loads, are not 

addressed herein. The reader is referred to the ASCE 7 standard and applicable building code 

provisions regarding special loads.

3.2 Load Combinations

The load combinations in table 3.1 are recommended for use with design 

specifications based on ASD and LRFD. Load combinations provide the basic set of 

building load conditions that should be considered by the designer. They establish the 

proportioning of multiple transient loads that may assume point-in-time values when 

the load of interest attains its extreme design value. Load combinations are intended 

as a guide to the designer, who should exercise judgment in any particular 

application. The load combinations in table 3.1 are appropriate for use with the design 

loads determined in accordance with this chapter.

The principle used to proportion loads is a recognition that when one load 

attains its maximum lifetime value, the other loads assume arbitrary point-in-time 

values associated with the structure’s normal or sustained loading conditions. The 

advent of LRFD has drawn greater attention to this principle (Ellingwood et al., 1982; 

Galambos et al., 1982). The proportioning of loads in this chapter for ASD is 

consistent with design load specifications such as ASCE 7. ASD load combinations

found in building codes typically have included some degree of proportioning (that is, 

D + W + 1/2S) and usually have made allowance for a special reduction for multiple 

transient loads. Some earlier codes also have permitted allowable material stress 

increases for load combinations involving wind and earthquake loads. None of these 

adjustments for ASD load combinations are recommended for use with table 3.1 

because the load proportioning is considered sufficient. However, allowable material 

stress increases that are based upon the duration of the load (that is, wood members 

under wind loading) may be combined with load proportioning.

Note also that the wind load factor of 1.0 in table 3.1 used for LRFD is 

consistent with current wind design practice and now recognizes ultimate wind loads 

when the speeds illustrated in the ASCE 7-10 maps are used. The return period of the 

design wind speeds for residential buildings along the hurricane-prone coast is now 
Residential Structural Design Guide 3-3

700 years, and this long return period provides a consistent risk basis for wind design 

across the country. Many elements of residential design continue to use ASD design 

level wind speeds, however, primarily because of how products have been tested,

rated, and marketed to the industry. Some prescriptive design documents such as the 

Wood Frame Construction Manual (WFCM) continue to use ASD load combinations 

in the development of loads provided in the design tables of that document (AWC, 

2012). The conversion of LRFD speeds to ASD speeds is ASD speed = LRFD 

speedx√0.6. The conversion of LRFD pressures to ASD pressures is ASD wind 

pressure = LRFD pressure x 0.6 (the ASD wind load factor). The load factor changes 

used in ASCE 7-10 are referenced in the 2012 editions of the building codes where 

ASCE 7-10 is referenced.

The load combinations in table 3.1 are simplified and tailored to specific 

application in residential construction and the design of typical components and 

systems in a home. These or similar load combinations often are used in practice as 

shortcuts to those load combinations that govern the design result. This guide makes 

effective use of the shortcuts and demonstrates them in the examples provided later in 

the chapter. The shortcuts are intended only for the design of residential light-frame 

construction.
Residential Structural Design Guide 3-4

TABLE 3.1 Typical Load Combinations Used for the Design of 

Components and Systems1

Component or System ASD Load Combinations LRFD Load Combinations

Foundation wall

(gravity and soil lateral loads)

D + H

D + H + 0.75 (Lr or S) + 0.75L2

1.2D + 1.6H

1.2D + 1.6H + 1.6L2 + 0.5(Lr + S)

1.2D + 1.6H + 1.6(Lr or S) + L2

Headers, girders, joists, interior loadbearing walls and columns, footings 

(gravity loads)

D + 0.75 L2 + 0.75 (Lr or S)

D +0.75 (Lr or S) + 0.75 L2

1.2D + 1.6L2 + 0.5 (Lr or S)

1.2D + 1.6(Lr or S) + L2

Exterior load-bearing walls and 

columns (gravity and transverse 

lateral load) 3

Same as immediately above, 

plus 0.6D + 0.6W

D + 0.7E + 0.75L2 + 0.75S4

Same as immediately above, plus

1.2D + 1.0W

1.2D + 1.0E + L2 + 0.2S4

Roof rafters, trusses, and beams; roof 

and wall sheathing (gravity and wind 

loads)

D + (Lr or S)

0.6D + 0.6Wu

5

0.6D + 0.6W

1.2D + 1.6(Lr or S)

0.9D + 1.0Wu

5

1.2D + 1.0W

Floor diaphragms and shear walls

(in-plane lateral and overturning 

loads) 6

0.6D + (0.6W or 0.7E) 0.9D + (1.0W or 1.0E)

Notes:

1

The load combinations and factors are intended to apply to nominal design loads defined as follows: D = estimated mean dead weight of the 

construction; E = design earthquake load; H = design lateral pressure for soil condition/type; L = design floor live load; Lr = maximum roof live 

load anticipated from construction/maintenance;; S = design roof snow load; and W = design wind load. The design or nominal loads should be 

determined in accordance with this chapter.

2

Attic loads may be included in the floor live load, but a 10-psf attic load typically is used only to size ceiling joists adequately for access 

purposes. If the attic is intended for storage, however, the attic live load (or some portion) should also be considered for the design of other 

elements in the load path.

3

The transverse wind load for stud design is based on a localized component and cladding wind pressure; D + W provides an adequate and 

simple design check representative of worst-case combined axial and transverse loading. Axial forces from snow loads and roof live loads 

should usually not be considered simultaneously with an extreme wind load because they are mutually exclusive on residential sloped roofs. 

Further, in most areas of the United States, design winds are produced by either hurricanes or thunderstorms; therefore, these wind events and 

snow are mutually exclusive because they occur at different times of the year.

4

For walls supporting heavy cladding loads (such as brick veneer), an analysis of earthquake lateral loads and combined axial loads should be 

considered; however, this load combination rarely governs the design of light-frame construction. 5

Wu is wind uplift load from negative (that is, suction) pressures on the roof. Wind uplift loads must be resisted by continuous load path 

connections to the foundation or until offset by D. 6

The 0.6 reduction factor on D is intended to apply to the calculation of net overturning stresses and forces. For wind, the analysis of overturning 

should also consider roof uplift forces unless a separate load path is designed to transfer those forces.

3.3 Dead Loads

Dead loads consist of the permanent construction material loads comprising the roof, floor, 

wall, and foundation systems, including claddings, finishes, and fixed equipment. The values 

for dead loads in table 3.2 are for commonly used materials and constructions in light-frame 

residential buildings. Dead loads are given as nominal or ASD-level loads. Table 3.3 provides 

values for common material densities and may be useful in calculating dead loads more 

accurately. The design examples in section 3.12 demonstrate the straightforward process of 

calculating dead loads.
Residential Structural Design Guide 3-5

TABLE 3.2 Dead Loads for Common Residential Construction1

Roof Construction

Light-frame wood roof with wood structural panel 

sheathing and 1/2-inch gypsum board ceiling (2 psf) with 

asphalt shingle roofing (3 psf)

- with conventional clay/tile roofing

- with lightweight tile

- with metal roofing

- with wood shakes

- with tar and gravel

15 psf

27 psf

21 psf

14 psf

15 psf

18 psf

Floor Construction

Light-frame 2x12 wood floor with 3/4-inch wood 

structural panel sheathing and 1/2-inch gypsum board 

ceiling (without 1/2-inch gypsum board, subtract 2 psf 

from all values) with carpet, vinyl, or similar floor 

covering

10 psf2

- with wood flooring 12 psf

- with ceramic tile 15 psf

- with slate 19 psf

Wall Construction

Light-frame 2x4 wood wall with 1/2-inch wood 

structural panel sheathing and 1/2-inch gypsum board 

finish (for 2x6, add 1 psf to all values)

6 psf

- with vinyl or aluminum siding 7 psf

- with lap wood siding 8 psf

- with 7/8-inch portland cement stucco siding 15 psf

- with thin-coat stucco on insulation board 9 psf

- with 3-1/2-inch brick veneer 45 psf

Interior partition walls (2x4 with 1/2-inch gypsum board 

applied to both sides) 6 psf

Foundation Construction

6-inch-thick wall

8-inch-thick wall

10-inch-thick wall

12-inch-thick wall

6-inch x 12-inch concrete footing

6-inch x 16-inch concrete footing

8-inch x 24-inch concrete footing

Masonry3

 Concrete

Hollow Solid or Full Grout

28 psf 60 psf 75 psf

36 psf 80 psf 100 psf

44 psf 100 psf 123 psf

50 psf 125 psf 145 psf

 73 plf

 97 plf

193 plf

psf = pounds per square foot

Notes:

1

For unit conversions, see appendix B.

2

Value also used for roof rafter construction (that is, cathedral ceiling). 3

For partially grouted masonry, interpolate between hollow and solid grout in accordance with the fraction of masonry cores that are grouted.
Residential Structural Design Guide 3-6

TABLE 3.3 Densities for Common Residential Construction Materials1

Aluminum

Copper

Steel

Concrete (normal weight with light reinforcement)

Masonry, grout

Masonry, brick

Masonry, concrete

Glass

Wood (approximately 10 percent moisture content)2

- spruce-pine-fir (G = 0.42)

- spruce-pine-fir, south (G = 0.36)

- southern yellow pine (G = 0.55)

- Douglas fir–larch (G = 0.5)

- hem-fir (G = 0.43)

- mixed oak (G = 0.68)

Water

Structural wood panels

- plywood

- oriented strand board

Gypsum board

Stone

- Granite

- Sandstone

Sand, dry

Gravel, dry 

170 pcf

556 pcf

492 pcf

145–150 pcf

140 pcf

100–130 pcf

85–135 pcf

160 pcf

29 pcf

25 pcf

38 pcf

34 pcf

30 pcf

47 pcf

62.4 pcf

36 pcf

36 pcf

50 pcf

96 pcf

82 pcf

90 pcf

104 pcf

pcf = pounds per cubic foot

Notes:

1

For unit conversions, see appendix B.

2

The equilibrium moisture content of lumber is usually not more than 10 percent in protected building construction. The specific gravity, G, is 

the decimal fraction of dry wood density relative to that of water; therefore, at a 10 percent moisture content, the density of wood is 1.1(G)(62.4 

lbs/ft3

). The values given are representative of average densities and may easily vary by as much as 15 percent, depending on lumber grade and 

other factors.

3.4 Live Loads

Live loads are produced by the use and occupancy of a building. Loads include those from 

human occupants, furnishings, nonfixed equipment, storage, and construction and 

maintenance activities. Table 3.4 provides recommended design live loads for residential 

buildings. Live loads also are given as nominal or ASD-level loads. Example 3.1 in section 

3.10 demonstrates use of those loads and the load combinations specified in table 3.1, along 
Residential Structural Design Guide 3-7

with other factors discussed in this section. As required to adequately define the loading 

condition, loads are presented in terms of uniform area loads (in pounds per square foot: psf), 

concentrated loads (in pounds: lbs), and uniform line loads (in pounds per linear foot: plf). 

The uniform and concentrated live loads should not be applied simultaneously in a structural 

evaluation. Concentrated loads should be applied to a small area or surface consistent with the 

application and should be located or directed to give the maximum load effect possible in enduse conditions. For example, the stair concentrated load of 300 pounds should be applied to 

the center of the stair tread between supports. The concentrated wheel load of a vehicle on a 

garage slab or floor should be applied to all areas or members subject to a wheel or jack load, 

typically using a loaded area of about 20 square inches.

TABLE 3.4 Live Loads for Residential Construction1

Application Uniform Load Concentrated Load

Roof2

Slope ≥ 4:12 15 psf 250 lbs

Flat to 4:12 slope 20 psf 250 lbs

Attic3

Without storage 10 psf 250 lbs

With storage 20 psf 250 lbs

Floors

Bedroom areas3,4 30 psf 300 lbs

Other areas 40 psf 300 lbs

Garages 50 psf 2,000 lbs (passenger cars, vans, 

light trucks)

Decks and balconies 40 psf7

Stairs 40 psf 300 lbs

Guards and handrails 50 plf5 200 lbs

 Guard in-fill components 50 psf6

Grab bars N/A 250 lbs

lbs = pounds; plf = pounds per linear foot; psf = pounds per square foot

Notes:

1

Live load values should be verified relative to the locally applicable building code.

2

Roof live loads are intended to provide a minimum load for roof design in consideration of maintenance and construction activities. They 

should not be considered in combination with other transient loads (for example, floor live load, wind load) when designing walls, floors, and 

foundations. A 15-psf roof live load is recommended for residential roof slopes greater than 4:12; refer to ASCE 7-10 for an alternate approach. 3

Loft sleeping and attic storage loads should be considered only in areas with a clear height greater than about 3.5 feet. The concept of a “clear 

height” limitation on live loads is logical, but it may not be universally recognized. 4

Some codes require 40 psf for all floor areas.

5 ASCE 7-10 indicates that this load does not have to be considered for one- and two-family dwellings. 6 The applied normal load on an area is not to exceed 12 in. by 12 in. 7 ASCE 7 requirements may be more stringent.

The floor live load on any given floor area may be reduced in accordance with equation 3.4-1 

(Harris, Corotis, and Bova, 1981). Live load reductions also are allowed for multiple floors in 

ASCE 7-10. The equation applies to floor and support members, such as beams or columns 

(see table 3-5), which experience floor loads from a total tributary floor area greater than 200 

square feet. This equation also is in chapter 4 of ASCE 7-10, which covers live load design.
Residential Structural Design Guide 3-8

Equation 3.4-1

where

L = reduced design live load per ft2 of area supported by the member

KLL = live load element factor

Lo = unreduced design live load per ft2 of area supported by the member

AT = the tributary area in ft2

L shall not be less than 0.50L0 for members supporting one floor and not less than 

0.40L0 for members supporting two or more floors.

TABLE 3.5 Live Load Element Factor, KLL

Element KLL*

Interior columns 4

Exterior columns without cantilever slabs 4

Edge columns with cantilever slabs 3

Corner columns with cantilever slabs 2

Edge beams without cantilever slabs 2

Interior beams 2

All other members not identified, including 1

Edge beams with cantilever slabs

Cantilever beams

One-way slabs

 Two-way slabs

 Members without provisions for continuous

 shear transfer normal to their span

*In lieu of the preceding values, KLL may be calculated.

Note also that the nominal design floor live load in table 3.4 includes both a 

sustained and a transient load component. The sustained component is that load 

typically present at any given time and includes the load associated with normal 

human occupancy and furnishings. For residential buildings, the mean sustained live 

load is about 6 psf but typically varies from 4 to 8 psf (Chalk and Corotis, 1978). The 

mean transient live load for dwellings also is about 6 psf but may be as high as 13 psf. 

A total design live load of 30 to 40 psf is therefore fairly conservative.

3.5 Soil Lateral Loads

The lateral pressure exerted by earth backfill against a residential foundation 

wall (basement wall) can be calculated with reasonable accuracy on the basis of 

theory, but only for conditions that rarely occur in practice (Peck, Hanson, and 

Thornburn, 1974; University of Alberta, 1992). Theoretical analyses usually are 

based on homogeneous materials that demonstrate consistent compaction and 

behavioral properties. Such conditions rarely are experienced in typical residential 

construction projects.
Residential Structural Design Guide 3-9

The most common method of determining lateral soil loads on residential 

foundations follows Rankine’s (1857) theory of earth pressure and uses what is 

known as the Equivalent Fluid Density (EFD) method. As shown in figure 3.1, 

pressure distribution is assumed to be triangular and to increase with depth.

In the EFD method, the soil unit weight w is multiplied by an empirical coefficient Ka to 

account for the soil is not actually fluid and the pressure distribution is not necessarily 

triangular. The coefficient Ka is known as the active Rankine pressure coefficient. The EFD is 

determined as shown in equation 3.5-1.

Equation 3.5-1 q = Kaw

FIGURE 3.1 Triangular Pressure Distribution on a Basement Foundation 

Wall

For the triangular pressure distribution shown in figure 3.1, the pressure, P in 

psf, at depth, h in feet, is determined by equation 3.5-2, and the resultant force, H in 

lbs, at depth, h in feet, is determined by equation 3.5-3. The factor q is the EFD as 

discussed above.
Residential Structural Design Guide 3-10

Equation 3.5-2

P = qh

The total active soil force (pounds per linear foot of wall length) is—

Equation 3.5-3 2 qh

2

1 (qh)(h) 2

1 H = =

where h = the depth of the unbalanced fill on a foundation wall

H = the resultant force (plf) applied at a height of h/3 from the base of the 

unbalanced fill because the pressure distribution is assumed to be triangular

The EFD method is subject to judgment as to the appropriate value of the 

coefficient Ka. The values of Ka in table 3.6 are recommended for the determination 

of lateral pressures on residential foundations for various types of backfill materials 

placed with light compaction and good drainage. Given the long-time use of a 30

pounds per cubic foot (pcf) EFD in residential foundation wall prescriptive design 

tables (ICC, 2012), the values in table 3.6 may be considered somewhat conservative 

for typical conditions. A relatively conservative safety factor of 3 to 4 is typically 

applied to the design of unreinforced or nominally reinforced masonry or concrete 

foundation walls (ACI, 2011). Therefore, at imminent failure of a foundation wall, the 

30 psf design EFD would correspond to an active soil lateral pressure determined by 

using an EFD of about 90 to 120 pcf or more. The design examples in chapter 4 

demonstrate the calculation of soil loads.

TABLE 3.6 Values of Ka , Soil Unit Weight, and Equivalent Fluid 

Density by Soil Type1,2,3

Type of Soil4

(Unified Soil Classification)

Active Pressure

Coefficient (Ka)

Soil Unit 

Weight (pcf)

Equivalent Fluid 

Density (pcf)

Sand or gravel (GW, GP, GM, SW, SP) 0.26 115 30

Silty sand, silt, and sandy silt (GC, SM) 0.35 100 35

Clay-silt, silty clay (SM-SC, SC, ML, MLCL)0.45 100 45

Clay5 (CL, MH, CH) 0.60 100 60

Notes:

1 Values are applicable to well-drained foundations with less than 10 feet of backfill placed with light compaction or natural settlement, as is 

common in residential construction. The values do not apply to foundation walls in flood-prone environments; in such cases, an equivalent fluid 

density value of 80 to 90 pcf would be more appropriate (HUD, 1977).

2 Values are based on the Standard Handbook for Civil Engineers, 3rd ed. (Merritt, 1983), and on research on soil pressures reported in Thin 

Wall Foundation Testing, Department of Civil Engineering, University of Alberta, (March 1992). The designer should note that the values for 

soil equivalent fluid density differ from those recommended in ASCE 7-10 but are nonetheless compatible with current residential building 

codes, design practice, and the stated references.

3 These values do not consider the significantly higher loads that can result from expansive clays and the lateral expansion of moist, frozen soil. 

Such conditions should be avoided by eliminating expansive clays adjacent to the foundation wall and providing for adequate surface and 

foundation drainage.

4 Organic silts and clays and expansive clays are unsuitable for backfill material. 5 Backfill in the form of clay soils (non-expansive) should be used with caution on foundation walls with unbalanced fill heights greater than 3 

to 4 feet and on cantilevered foundation walls with unbalanced fill heights greater than 2 to 3 feet.
Residential Structural Design Guide 3-11

Depending on the type and depth of backfill material and the manner of its 

placement (see table 3.7), common practice in residential construction is to allow the 

backfill soil to consolidate naturally by providing an additional 3 to 6 inches of fill 

material. The additional backfill ensures that surface water drainage away from the 

foundation remains adequate (that is, the grade slopes away from the building). It also 

helps avoid heavy compaction that could cause undesirable loads on the foundation 

wall during and after construction. If soils are heavily compacted at the ground 

surface or compacted in lifts to standard Proctor densities greater than approximately

85 percent of optimum (ASTM, 2012), the standard 30 pcf EFD assumption may be 

inadequate. In cases in which the backfill supports exterior slabs, patios, stairs, or 

other items, however, some amount of compaction is required unless the structures 

are supported on a separate foundation bearing on undisturbed ground.

Some remediation may be necessary in areas that contain marine clay or other expansive soils. 

In very moist conditions, these soils can place significant lateral loads against foundation 

walls. The soils may need to be replaced with soil of lower clay content or the moisture levels 

must be stabilized to reduce excessive lateral pressures.

TABLE 3.7 Lateral Soil Load

Description of Backfill Material3

Unified Soil 

Classification

Design Lateral Soil Load1 (pound 

per square foot per foot of depth)

Active Pressure At-Rest Pressure

Well-graded, clean gravels; gravel-sand mixes GW 30 60

Poorly graded clean gravels; gravel-sand mixes GP 30 60

Silty gravels; poorly graded gravel-sand mixes GM 40 60

Clayey gravels; poorly graded gravel-clay mixes GC 45 60

Well-graded, clean sands; gravelly sand mixes SW 30 60

Poorly graded clean sands; sand-gravel mixes SP 30 60

Silty sands; poorly graded sand-silt mixes SM 45 60

Sand-silt clay mix with plastic fines SM-SC 45 100

Clayey sands; poorly graded sand-clay mixes SC 60 100

Inorganic silts; clayey silts ML 45 100

Inorganic silt-clay mixes ML-CL 60 100

Inorganic clays of low to medium plasticity CL 60 100

Organic silts and silt clays of low plasticity OL 2 2

Inorganic clayey silts; elastic silts MH 2 2

Inorganic clays of high plasticity CH 2 2

Notes:

1 Design lateral soil loads are given for moist conditions for the specified soils at their optimum densities. Actual field conditions shall govern. 2 Unsuitable as backfill material 3 The definition and classification of soil materials is in accordance with ASTM D2487.
Residential Structural Design Guide 3-12

3.6 Wind Loads

3.6.1 General

Wind produces dynamic loads on a structure at highly variable magnitudes. 

The variation in pressures at different locations on a building is complex to the point 

that pressures may become too analytically intensive for precise consideration in 

design. Wind load specifications attempt to simplify the design problem by 

considering basic static pressure zones on a building representative of peak loads that 

are likely to be experienced. The peak pressures in one zone for a given wind 

direction may not, however, occur simultaneously with peak pressures in other zones. 

For some pressure zones, the peak pressure depends on a narrow range of wind 

directions; therefore, the wind directionality effect must also be factored into 

determining risk-consistent wind loads on buildings. Characteristics of the building 

site and the surrounding area, such as exposure and topography, also play a large role 

in determining the peak pressures on the structure and should be carefully considered. 

In fact, most modern wind load specifications account for wind directionality and 

other effects in determining nominal design loads in some simplified form (ASCE, 

2010). This section further simplifies wind load design specifications to provide an 

easy yet effective approach for designing typical residential buildings.

Because they vary substantially over the surface of a building, wind loads are 

considered at two different scales. On a large scale, the loads produced on the overall 

building are resisted by a system of structural elements working together to transfer 

the wind loads acting on the entire structure to the ground, a system known as the 

main wind force-resisting system (MWFRS). The MWFRS of a home includes the 

shear walls and diaphragms that create the lateral force-resisting system (LFRS) as 

well as the structural systems, such as trusses, that experience loads from external and 

internal pressures generated on the building. The wind loads applied to the MWFRS 

account for the area-averaging effects of time-varying wind pressures on the surface 

or surfaces of the building.

Wind pressures are greater on certain localized surface areas of the building, 

particularly near abrupt changes in building geometry (for example, eaves, ridges, and 

corners). Those higher wind pressures can occur on smaller areas, particularly 

affecting the loads carried by components and cladding (for example, sheathing, 

windows, doors, purlins, and studs). The components and cladding (C&C) transfer 

localized time-varying loads to the MWFRS, at which point the loads average out 

both spatially and temporally since, at a given time, some components may be at near 

peak loads while others are at substantially less than peak.

In light-framed wood structural systems, the distinction between MWFRS and 

C&C is not as clear-cut as in other buildings. In some cases, structural components 

may act as MWFRS and as C&C, depending on situations. The designer must 

consider which elements of the building must be treated as C&C, part of the 

MWFRS, or both. As indicated, parts of the MWFRS that collect and transfer lateral 

loads in shear walls and floors or roof diaphragms consist of wall studs, sheathing, 
Residential Structural Design Guide 3-13

and trusses, and these elements as a system must be designed for MWFRS lateral 

loads; but the studs, sheathing, and truss chords must be designed for the direct 

loading from wind as C&C. Thus, the stud size and connection to top and bottom 

plates must be designed for C&C pressures, yet the entire wall system, especially the 

sheathing thickness and the nailing attachment of the sheathing to the studs, must be 

designed to resist the shear forces created by the lateral loads.

The next section presents a simplified method for determining both MWFRS 

and C&C wind loads. Because the loads in section 3.6.2 are determined for specific 

applications, the calculation of MWFRS and C&C wind loads is implicit in the values 

provided. Design example 3.2 in section 3.12 demonstrates the calculation of wind 

loads by applying the simplified method in the following section to several design 

conditions associated with wind loads and the load combinations presented in table 

3.1.

3.6.2 Determination of Wind Loads on

Residential Buildings

The following method for the design of residential buildings is based on a 

simplification of the ASCE 7-10 wind provisions (ASCE, 2010); therefore, the wind 

loads are not exact duplicates. Lateral loads and roof uplift loads are determined by 

using a projected area approach. Other wind loads are determined for specific 

components or assemblies that comprise the exterior building envelope. Determining

design wind loads on a residential building and its components requires five steps.

Step 1: Determine site design wind speed and basic velocity pressure

From the wind map in figure 3.2 (refer to ASCE 7-10 for a more detailed map for risk 

category II buildings), select a design wind speed for the site (ASCE, 2010), or, alternatively, 

find a location-specific wind speed from the local building code office or by using 

www.atcouncil.org/windspeed. The wind speed map in ASCE 7-10 (figure 3.2) includes the 

most accurate data and analysis available regarding design wind speeds in the United States. 

The ASCE 7-10 wind speeds are higher than those used in older design wind maps. The 

difference results solely from using ultimate wind speeds developed for use with 700-year 

return periods for risk category II buildings that include residential uses. The speeds 

correspond to approximately a 7 percent probability of exceedance in 50 years. The design 3-

second peak gust wind speeds are 110 to 115 miles per hour (mph) in most of the United 

States; however, along the hurricane-prone Gulf and Atlantic coasts, the design wind speeds 

range from 115 to 180 mph. The wind speeds are standardized for exposure C conditions at 33 

feet (10 meters). Tornadoes have not been considered in the design wind speeds presented in 

figure 3.2. Design loads for tornadoes are still in the development stage, and discussion of the 

latest knowledge is provided in section 3.10.

Once the nominal design wind speed in terms of peak 3-second gust is 

determined, the designer can select the basic velocity pressure, in accordance with 

table 3.8. The basic velocity pressure is a reference wind pressure to which 

coefficients are applied to determine the surface pressures on a building. Velocity 

pressures in table 3.8 are based on typical conditions for residential construction, 
Residential Structural Design Guide 3-14

namely, suburban terrain (exposure B) and relatively flat or rolling terrain without 

topographic wind speed-up effects.

FIGURE 3.2a Basic Design Wind Speed Map from ASCE 7-10

Source: Reprinted with permission from the American Society of Civil Engineers, 1801 Alexander Bell Road, Reston, VA. Copyright ASCE.
Residential Structural Design Guide 3-15

FIGURE 3.2b Basic Design Wind Speed Map from ASCE 7-10

Source: Reprinted with permission from the American Society of Civil Engineers, 1801 Alexander Bell Road, Reston, VA. Copyright ASCE.

Location V mph (m/s)

Guam 195 (87)

Virgin Islands 165 (74)

American Samoa 160 (72)

Hawaii – Special Wind Region Statewide 130 (58)

Notes:

1. Values are design 3-second gust wind speed in miles per hour (m/s) at 33 ft (10m) above ground for 

Exposure C category.

2. Linear interpolation between contours is permitted.

3. Islands and coastal areas outside the last contour shall use the last wind speed contour of the coastal area.

4. Mountainous terrain, gorges, ocean promontories, and special wind regions shall be examined for unusual 

wind conditions.

5. Wind speeds correspond to approximately a 7% probability of exceedance in 50 years (annual exceedance 

probability = 0.00143, MRI 700 years).
Residential Structural Design Guide 3-16

TABLE 3.8 Basic Wind Velocity Pressures (psf) for Suburban Terrain1

(MWFRS)

Design Wind Speed, V

(mph, peak gust)

One-Story Building 

(15')

(KZ = 0.57)2

Two-Story Building

(30')

(KZ = 0.7)2

Three-Story Building

(45')

(KZ = 0.78)

110

115

120

130

140

150

160

170

180

15

16

18

21

24

28

32

36

40

18

20

22

26

30

34

39

44

49

21

22

24

29

33

38

43

49

55

mph = miles per hour; MWFRS = main wind force-resisting system; psf = pounds per square foot.

Notes:

1

Velocity pressure (psf) equals 0.00256 KD KZV2, where KZ is the velocity pressure exposure coefficient associated with the vertical wind speed 

profile in suburban terrain (exposure B) at the mean roof height of the building. KD is the wind directionality factor, with a default value of 0.85. 

All pressures have been rounded to nearest whole psf.

2 To be compliant with ASCE 7-10, a minimum Kz of 0.7 should be applied to determine velocity pressure for one- and two-story buildings in 

exposure B (suburban terrain) for the design of components and cladding, in exposure B when the envelope procedure is used for the MWFRS, 

or when designing components and cladding.

Step 2: Adjustments to the basic velocity pressure

If appropriate, the basic velocity pressure from step 1 should be adjusted in 

accordance with the factors below. The adjustments are cumulative.

Open exposure. The wind pressure values in table 3.8 are based on typical 

residential exposures to wind (exposure B). If a site is located in generally open, flat 

terrain with few obstructions to the wind in most directions (exposure C), the designer 

should multiply the values in table 3.8 by a factor of 1.4. Exposure to a body of water 

(that is, an ocean or lake) increases wind pressures more because of reduced friction 

at the surface (exposure D). The values in table 3.8 should be multiplied by a factor of 

1.7 to account for this increased pressure for exposure D conditions. The factor may 

be used to adjust wind loads according to the exposure related to the specific 

directions of wind approach to the building. The wind exposure conditions used in 

this guide are derived from ASCE 7-10, and more information about how to 

determine these exposures is provided in the ASCE 7-10 commentary.

Wind directionality. As noted, the direction of the wind in a given event does 

not create peak loads (which provide the basis for design pressure coefficients) 

simultaneously on all building surfaces. In some cases, the pressure zones with the 

highest design pressures are extremely sensitive to wind direction. In accordance with 

ASCE 7-10, the velocity pressures in table 3.8 are based on a directionality 

adjustment of 0.85.

Topographic effects. If topographic wind speed-up effects are likely because a 

structure is located near the crest of a protruding hill or cliff, the designer should 

consider using the topographic factor provided in ASCE 7-10. Wind loads can be 

increased for buildings sited in particularly vulnerable locations relative to 

topographic features that cause localized wind speed-up for specific wind directions. 
Residential Structural Design Guide 3-17

The International Residential Code (IRC; ICC 2011) provides a “Simplified

Topographic Wind Speed-up Method” for the Kzt factor where required. The 

simplified method in the IRC is based on the wind speed-up effect for cliff edges, the 

most vulnerable of the three types of features (hills, ridges or escarpments), and on 

certain terrain feature heights and dwelling locations. If a more accurate and 

potentially less conservative determination of an adjusted design wind speed is 

desired, the designer can apply the ASCE 7-10 provisions for adjusting the wind 

speed to account for the Kzt factor, where required.

Step 3: Determine lateral wind pressure coefficients

Lateral pressure coefficients in table 3.9 are composite pressure coefficients 

that combine the effect of positive pressures on the windward face of the building and 

negative (suction) pressures on the leeward faces of the building. The lateral pressure 

coefficients are the total effect of the shape factor (Cp) and the gust effect factor (G). 

When multiplied by the velocity pressure from steps 1 and 2, the selected pressure 

coefficient provides a single wind pressure that is applied to the vertical projected 

area of the roof and wall, as indicated in table 3.9. The resulting load is then used to 

design the home’s LFRS (see chapter 6). The lateral wind load must be determined 

for the two orthogonal directions on the building (that is, parallel to the ridge and 

perpendicular to the ridge), using the vertical projected area of the building for each 

direction. Lateral loads are then assigned to various systems (for example, shear 

walls, floor diaphragms, and roof diaphragms) by use of tributary areas or other 

methods described in chapter 6.

This method can be used for determining shear loads because the internal pressures in the 

building cancel out and do not affect the shear loads. Overturning moments and the design of 

wall studs and lateral out-of-plane wall loads at the roof-to-wall connection must consider the 

effects of internal pressure, however; thus, the projected area method is not useful for those 

calculations. See step 4 for additional information.

TABLE 3.9 Lateral Pressure Coefficients for Application

to Vertical Projected Areas

Application Lateral Pressure Coefficients

Roof Vertical Projected Area (by slope)

Flat

3:12

6:12

≥9:12

0.0

0.43

0.77

0.85

Wall Projected Area 1.1
Residential Structural Design Guide 3-18

Step 4: Determine wind pressure coefficients for components

and assemblies

The pressure coefficients in table 3.9 are derived from ASCE 7-10, based on 

the assumption that the building is enclosed and not subject to higher internal 

pressures that may result from a windward opening in the building. Using the values 

in table 3.9 greatly simplifies the more detailed methodology described in ASCE 7-

10; as a result, some numbers are “rounded.” With the exception of the roof uplift 

coefficient, all pressures calculated with the coefficients are intended to be applied 

perpendicular to the building surface area that is tributary to the element of concern; 

thus, the wind load is applied perpendicular to the actual building surface, not to a 

projected area. The roof uplift pressure coefficient is used to determine a single wind 

pressure that may be applied to a horizontal projected area of the roof to determine 

roof tie-down connection forces.

For buildings in hurricane-prone regions subject to wind-borne debris, the 

GCp values in table 3.10 are still valid, but the glazed openings in the building must

be protected from the possibility of damage by wind-borne debris breaching a wall or 

roof opening, such as a window or skylight. Past versions of ASCE 7 had allowed 

design for a “partially enclosed” condition using higher internal pressure coefficients 

in wind-borne debris regions, but this technique allows a potentially significant 

amount of wind-driven rain into the building, which would still create a near total 

economic loss. ASCE 7 no longer allows this design method.

Step 5: Determine design wind pressures

Once the basic velocity pressure is determined in step 1 and adjusted in step 2 for exposure 

and other site-specific considerations, the designer can calculate the design wind pressures by 

multiplying the adjusted basic velocity pressure by the pressure coefficients selected in steps 3 

and 4. The lateral pressures on the MWFRS are based on coefficients from step 3 and are 

applied to the tributary areas of the LFRS, such as shear walls and diaphragms. The pressures 

based on coefficients from step 4 are applied to tributary areas of members such as studs, 

rafters, trusses, and sheathing to determine stresses in members and forces in connections.
Residential Structural Design Guide 3-19

TABLE 3.10 Wind Pressure Coefficients for Systems and Components 

(enclosed building)1

Application Pressure Coefficients (GCp)

2

Roof

Trusses, roof beams, ridge and hip/valley rafters 

Rafters and truss panel members 

Roof sheathing 

Skylights and glazing 

Roof uplift3

- hip roof with slope between 3:12 and 6:12

- hip roof with slope greater than 6:12

- all other roof types and slopes

Windward overhang4

-0.9, +0.4

-1.2, +0.7

-2.8, +0.7

-1.2, +1.0

-0.9

-0.8

-1.2

+ 0.7

Wall

All framing members 

Wall sheathing and cladding/siding

Windows, doors, and glazing

Garage doors

Air-permeable claddings5

-1.5, +1.1

-1.6, +1.2

-1.3, +1.2

-1.1, +1.0

-0.9, 0.8

Notes:

1

All coefficients include internal pressure in accordance with the assumption of an enclosed building. With the exception of the categories 

labeled trusses, roof beams, ridge and hip/valley rafters, and roof uplift, which are based on MWFRS loads, all coefficients are based on 

component and cladding wind loads.

2

Positive and negative signs represent pressures acting inwardly and outwardly, respectively, from the building surface. A negative pressure is a 

suction or vacuum. Both pressure conditions should be considered to determine the controlling design criteria.

3

The roof uplift pressure coefficient is used to determine uplift pressures that are applied to the horizontal projected area of the roof for the 

purpose of determining uplift tie-down forces. Additional uplift force on roof tie-downs resulting from roof overhangs should also be included. 

The uplift force must be transferred to the foundation or to a point where it is adequately resisted by the dead load of the building and the 

capacity of conventional framing connections.

4

The windward overhang pressure coefficient is applied to the underside of a windward roof overhang and acts upwardly on the bottom surface 

of the roof overhang. If the bottom surface of the roof overhang is the roof sheathing, or if the soffit is not covered with a structural material on 

its underside, then the overhang pressure shall be considered additive to the roof sheathing pressure. 5

Air-permeable claddings allow for pressure relief such that the cladding experiences about two-thirds of the pressure differential experienced 

across the wall assembly (FPL, 1999). Products that experience reduced pressure include lap-type sidings such as wood, vinyl, aluminum, and 

other similar sidings. Since these components are usually considered “nonessential,” it may be practical to multiply the calculated wind load on 

any nonstructural cladding by 0.75 to adjust for a serviceability wind load (Galambos and Ellingwood, 1986). Such an adjustment would also be 

applicable to deflection checks, if required, for other components listed in the table. However, a serviceability load criterion is not included or 

clearly defined in existing design codes.

3.6.3 Special Considerations in

Hurricane-Prone Environments

3.6.3.1 Wind-Borne Debris

The wind loads determined in the previous section assume an enclosed 

building. If glazing in windows and doors is not protected from wind-borne debris or 

otherwise designed to resist potential impacts during a major hurricane, a building is 

more susceptible to structural damage resulting from higher internal pressures that 
Residential Structural Design Guide 3-20

may develop with a windward opening. The potential for water damage to building 

contents also increases. Openings created in the building envelope during a major 

hurricane or tornado often are related to unprotected glazing, improperly fastened 

sheathing, or weak garage doors and their attachment to the building. Section 3.10 

discusses tornado design conditions.

In recent years, much attention has been focused on wind-borne debris, based on the results of 

many damage investigations. Little research has been done to quantify the magnitude or type 

of debris. The current wind-borne debris protection trigger is wind speed, and the requirement 

for wind-borne debris protection is all or nothing—meaning that, in accordance with ASCE 7-

10, protection must be provided where design wind speeds are 130 mph and the building is 

within one mile of the coastal mean high-water line, or anywhere the design wind speed is 

140 mph or greater. Conventional practice for wind-borne debris protection in residential 

construction usually is either impact-resistant shutters installed over glazed openings or 

impact-resistant glazing. The IRC still permits the use of wood structural panels (plywood or 

oriented strand board [OSB]) as opening protection for glazing in one- and two-story 

buildings to resist impacts from wind-borne debris. To use wood structural panels for opening 

protection, however, attachment hardware is required, with anchors permanently installed on 

the building. Impact-resistant glazing or protective devices must be tested using an approved 

test method, such as ASTM E1886 (ASTM, 2005) and ASTM E1996 (ASTM, 2009b).

Just what defines impact resistance and the level of impact risk during a hurricane continues 

to be the subject of much debate. Surveys of damage following major hurricanes have 

identified several factors that affect the level of debris impact risk, including

• wind climate (design wind speed);

• exposure (for example, suburban, wooded, height of surrounding 

buildings);

• development density (that is, distance between buildings);

• construction characteristics (for example, type of roofing, degree of wind 

resistance); and

• debris sources (for example, roofing, fencing, and gravel).

Current standards for selecting impact criteria for wind-borne debris protection do not 

explicitly consider all of those factors. Further, the primary debris source in typical residential 

developments is asphalt roof shingles, clay roof tiles, landscaping materials and driveway 

gravel, vinyl siding, and vegetation from trees and shrubs, some of which are not represented 

in existing impact test methods. Recent research has provided insight into performance 

expectations (Fernandez, Masters, and Gurley, 2010; Masters et al., 2010). These factors have 

a dramatic effect on the level of wind-borne debris risk. Table 3.11 presents an example of 

missile types used for current impact tests. Additional factors to consider include emergency 

egress or access in the event of fire when impact-resistant glazing or fixed shutter systems are 

specified, potential injury or misapplication during installation of temporary methods of 

window protection, and durability of protective devices and connection details (including 

installation quality) such that they themselves do not become a debris hazard over time.
Residential Structural Design Guide 3-21

TABLE 3.11 Missile Types for Wind-Borne Debris Impact Tests1,2

Description Velocity Energy

2 gram steel balls 130 fps 10 ft-lb

4.5 lb 2x4 40 fps 100 ft-lb

9.0 lb 2x4 50 fps 350 ft-lb

fps = feet per second; ft-lb = foot-pounds; lb = pounds

1

Consult ASTM E1886 (ASTM, 2005) for guidance on testing apparatus and methodology. 2

These missile types are not necessarily representative of the predominant types or sources of debris at any particular site. 

The steel balls are intended to represent small gravel that would be commonly used for roof ballast. The 2x4 missiles are 

intended to represent a direct, end-on blow from construction debris.

Homes that experience wind-borne debris damage may exhibit more catastrophic failures, 

such as a roof blowoff, but usually this occurs only when large elements of the building 

envelope fail, such as large windows or garage doors. Wind pressure can also cause failures in 

these large elements; therefore, in hurricane-prone regions, large windows and garage doors 

should be specified that meet both wind pressure and wind-borne debris impact requirements, 

and the attachment of those elements to structural framing should be carefully designed.

One additional element that requires consideration, and for which research is being conducted, 

is wind-driven rain. Most window manufacturers have products tested to some limitation on 

water infiltration, and under normal weather conditions those limitations usually are sufficient 

(usually up to 15 percent of the design wind pressure). Hurricane-force winds will drive rain 

horizontally and that water can penetrate between window units, under doors, and into soffits 

and other small places such that, even with attention to this issue, the water can cause 

significant damage. Both the designer and the builder must pay attention to the construction 

details at every building joint and every hole in the building envelope to ensure that water 

penetration during high winds is minimized (Salzano, Masters, and Katsaros, 2010).

3.6.3.2 Tips to Improve Performance

The following design and construction tips are simple considerations for 

reducing a building’s vulnerability to hurricane wind damage:

• One-story buildings are less vulnerable than two- or three-story buildings

to wind damage.

• On average, hip roofs have demonstrated better performance than gableend roofs.

• Moderate roof slopes (that is, 5:12 to 6:12) tend to optimize the tradeoff

between lateral loads and roof uplift loads (that is, they are 

aerodynamically efficient).

• Roof sheathing installation should be inspected for the proper type and 

spacing of fasteners, particularly at connections to gable-end framing.

• The installation of metal strapping or other tie-down hardware should be 

inspected as required to ensure the transfer of uplift loads.

• If composition roof shingles are used, the shingles should be tested in 

accordance with ASTM D7158 (ASTM, 2011). All roof coverings should 

be designed or tested and installed to resist the applicable wind loads.
Residential Structural Design Guide 3-22

• Glazed-opening protection should be considered in the most severe 

hurricane-prone areas and in those areas defined as requiring wind-borne 

debris protection.

• The roof deck may be sealed or a secondary water barrier may be 

installed on the roof to prevent water infiltration in the event the primary 

roof covering is blown off. A sealed roof deck can be created by 

installing minimum 4-inch-wide strips of self-adhering underlayment 

complying with ASTM D1970 over the roof sheathing joints (ASTM, 

2009c). The IRC also contains enhanced underlayment specifications for 

high-wind regions that require the use of ASTM D226 Type II (30 pound) 

or equivalent underlayment with a rigorous fastening schedule (ASTM, 

2009a).

The HUD document Safer, Stronger Homes (HUD, 2011) includes further 

details regarding methods for improving the wind hazard resilience of new and 

existing residential structures.

3.7 Snow Loads

For design purposes, snow typically is treated as a simple uniform gravity 

load on the horizontal projected area of a roof. The uniformly distributed design snow 

load on residential roofs can be easily determined by using the unadjusted ground 

snow load. This simple approach also represents standard practice in some regions of 

the United States; however, it does not account for a reduction in roof snow load that 

may be associated with steep roof slopes with slippery surfaces (refer to ASCE 7-10). 

Drift loads on sloped gable or hip roofs must consider roof slope, warm and cold roof 

slope factors, and ridge-to-eave distances. ASCE 7-10 has design parameters for each 

of these snow and roof conditions. Drifting snow has caused numerous roof failures 

in the past 5 to 10 years. The design guidance in ASCE 7 addresses some of the issues 

important to consider for drifting snow; for building design, drifting snow must be 

considered at any building intersection where a roof adjoins a wall or other vertical 

surface where snow can accumulate. For buildings, snow drifting can occur where a 

building extension such as a garage or first floor addition adjoins an existing twostory wall. The problem of loading is complicated because snow loads vary with 

moisture content as well as depth, and depths vary with roof slope, wind speed, and

vertical height where drifting can occur.

Design ground snow loads may be obtained from the map in figure 3.3 (for a 

larger ground snow load map with greater detail, refer to ASCE 7-10); however, snow 

loads usually are defined by the local building department. Typical ground snow 

loads range from 0 psf in the southern United States to 50 psf in the northern United 

States. In mountainous areas, the ground snow load can surpass 100 psf, so local 

snow data must be carefully considered. The ASCE 7-10 map includes varying 

ground snow loads with ground elevation above sea level. In areas where the ground 

snow load is less than 15 psf, the minimum roof live load (refer to section 3.4) usually 

is the controlling gravity load in roof design.
Residential Structural Design Guide 3-23

FIGURE 3.3a Ground Snow Loads (ASCE 7-10) 

Source: Reprinted with permission from the American Society of Civil Engineers, 1801 Alexander Bell Road, Reston, VA. Copyright ASCE.

In CS areas, site-specific Case Studies are required to establish ground snow 

loads. Extreme local variations in ground snow loads in these areas 

preclude mapping at this scale.

Numbers in parentheses represent the upper elevation limits in feet for the 

ground snow load values presented below. Site-specific case studies are 

required to establish ground snow loads at elevations not covered.

To convert lb/sq ft to kN/m2

, multiply by 0.0479

To convert feet to meters, multiply by 0.3048
Residential Structural Design Guide 3-24

FIGURE 3.3b Ground Snow Loads (ASCE 7-10) 

Source: Reprinted with permission from the American Society of Civil Engineers, 1801 Alexander Bell Road, Reston, VA. Copyright ASCE.
Residential Structural Design Guide 3-25

3.8 Earthquake Loads

3.8.1 General

This section provides a simplified earthquake load analysis procedure 

appropriate for use in residential light-frame construction of not more than three 

stories above grade. As described in chapter 2, the lateral forces associated with 

seismic ground motion are based on fundamental Newtonian mechanics (F = ma), 

expressed in terms of an equivalent static load. The method provided in this section is 

a simplification of the most current seismic design provisions NEHRP (National 

Earthquake Hazards Reduction Program) Recommended Seismic Provisions for New 

Buildings and Other Structures (FEMA, 2009a). The method herein also is similar to 

a simplified approach found in more recent building codes (ICC, 2012).

In general, wood-framed homes have performed well from a life safety 

standpoint in major seismic events, probably because of, among other factors, their 

lightweight and resilient construction, the strength provided by nonstructural systems 

such as interior walls, and their load distribution capabilities. Wood-framed homes 

have not performed as well from a damage-reduction standpoint, in part because of 

brittle finishes such as gypsum board and masonry exteriors, as well as insufficient 

anchorage of wall framing to foundations, lack of sheathing on cripple walls, and 

slope failures on hillside sites (HUD, 1994). Garages with wide doors or houses with 

many large windows on the ground floor can fail as a result of so-called “soft story” 

or “weak story” behavior because the garage or ground floor walls are much less stiff 

than the roof or stories above.
Residential Structural Design Guide 3-26

FIGURE 3.4a Mapped Risk-Targeted Maximum Considered Earthquake 

(MCER) Spectral Response Acceleration

Source: Reprinted with permission from the American Society of Civil Engineers, 1801 Alexander Bell Road, Reston, VA. Copyright ASCE.

DISCUSSION

Maps prepared by United States Geological Survey (USGS) in 

collaboration with the Federal Emergency Management Agency 

(FEMA)-funded Building Seismic Safety Council (BSSC) and the 

American Society of Civil Engineers (ASCE). The basis is 

explained in commentaries prepared by BSSC and ASCE and in the 

references.

Ground motion values contoured on these maps incorporate:

• A target risk of structural collapse equal to 1% in 50 years based 

upon a generic structural fragility

• A factor of 1.1 to adjust from a geometric mean to the maximum 

response regardless of direction

• Deterministic upper limits imposed near large, active faults, 

which are taken as 1.8 times the estimated median response to the 

characteristic earthquake for the fault (1.8 is used to represent the 

84th percentile response), but not less than 150% g.

As such, the values are different from those on the uniform 

hazard USGS National Seismic Hazard Maps posted at: 

http://earthquake.usgs.gov/hazmaps.

Larger, more detailed versions of these maps are not provided 

because it is recommended that the corresponding USGS web tool 

(http://earthquake.usgs.gov/designmaps or 

http://content.seinstitute.org) be used to determine the mapped 

value for a specified location.

REFERENCES

Building Seismic Safety Council, 2009, NEHRP 

Recommended Seismic Provisions for New 

Buildings and Other Structures: FEMA P-750/2009 

Edition, Federal Emergency Management Agency, 

Washington, DC.

Huang, Yin-Nan, Whittaker, A.S., and Luco, Nicolas, 

2008, Maximum spectral demands in the near-fault 

region, Earthquake Spectra, Volume 24, Issue 1, 

pp. 319-341.

Luco, Nicolas, Ellingwood, B.R., Hamburger, R.O., 

Hooper, J.D., Kimball, J.K., and Kircher, C.A., 

2007, Risk-Targeted versus Current Seismic 

Design Maps for the Coterminous United States, 

Structural Engineers Association of California 

2007 Convention Proceedings, pp. 163-175.

Petersen, M.D., Frankel, A.D., Harmsen, S.C., 

Mueller, C.S., Haller, K.M., Wheeler, R.L. 

Wesson, R.L., Zeng, Yuehua, Boyd, O.S., Perkins, 

D.M., Luco, Nicolas, Field, E.H., Wills, C.J., and 

Rukstales, K.S, 2008 Documentation for the 2008 

Update of the United States Seismic Hazard Maps: 

U.S. Geological Survey Open-File Report 2008-

1128, 61 p.
Residential Structural Design Guide 3-27

FIGURE 3.4b Mapped Risk-Targeted Maximum Considered Earthquake 

(MCER) Spectral Response Acceleration

Source: Reprinted with permission from the American Society of Civil Engineers, 1801 Alexander Bell Road, Reston, VA. Copyright ASCE.
Residential Structural Design Guide 3-28

3.8.2 Determination of Earthquake Loads on Houses

The total lateral force at the base of a building is called seismic base shear. The lateral force 

experienced at a particular story level is called the story shear. The story shear is greatest in 

the ground story and least in the top story. Seismic base shear and story shear (V) are 

determined in accordance with the following equation:

Equation 3.8-1

W

I

R

S V

e

DS















 = ,

where

SDS = the design spectral response acceleration in the short-period range 

determined by equation 3.8-2 (g)

R = THE RESPONSE MODIFICATION FACTOR (DIMENSIONLESS)

W = the effective seismic weight of the building or supported by the story 

under consideration (lb); 20 percent of the roof snow load is also included 

where the ground snow load exceeds 30 psf 

Ie = importance factor, which is 1.0 for residential buildings.

When determining story shear for a given story, the designer attributes to that 

story one-half of the dead load of the walls on the story under consideration and the 

dead load supported by the story (dead loads used in determining seismic story shear 

or base shear are found in section 3.3). For housing, the interior partition wall dead 

load is reasonably accounted for by the use of a 6 psf load distributed uniformly over 

the floor area. When applicable, the snow load may be determined in accordance with 

section 3.7. The inclusion of any snow load, however, is based on the assumption that 

the snow is always frozen solid and adhered to the building such that it is part of the 

building mass during the entire seismic event.

The design spectral response acceleration for short-period ground motion SDS

typically is used because light-frame buildings such as houses are believed to have a 

short period of vibration in response to seismic ground motion (that is, high natural 

frequency). For example, the building tested as part of the NEESWood project in 

2006 (Filiatrault et al., 2010) had an elastic period of 0.21 seconds, consistent with 

the 0.2-second period used to establish the short-period ground motions.

Values of SMS are from figure 3.4. For a larger map with greater detail, refer 

to ASCE 7-10 or find the response accelerations using the U.S. Geological Survey 

(USGS) seismic design maps, based on either latitude and longitude or zip codes: 

http://earthquake.usgs.gov/hazards/designmaps/usdesign.php. The value of SDS

should be determined in consideration of the mapped short-period spectral response 

acceleration SMS and the required soil site amplification factor Fa as follows:
Residential Structural Design Guide 3-29

Equation 3.8-2

2 / 3( )( ) DS MS Fa S = S

The value of SMS ranges from practically zero in low-risk areas to 3g in the 

highest risk regions of the United States. A typical value in high seismic areas is 1.5g. 

In general, wind loads control the design of the LFRS of light-frame houses when SMS

is low. 

Table 3.12 provides the values of Fa associated with a standard “firm” soil 

condition used for the design of residential buildings. Fa decreases with increasing 

ground motion because the soil begins to dampen the ground motion as shaking 

intensifies. The soil can therefore have a moderating effect on the seismic shear loads 

experienced by buildings in high-seismic-risk regions. Dampening also occurs 

between a building foundation and the soil and thus has a moderating effect. The soilstructure interaction effects on residential buildings have been the topic of little study; 

therefore, precise design procedures have yet to be developed. If a site is located on 

fill soils or “soft” ground, a different value of Fa should be considered (see ASCE, 

2010, for the full table). Nonetheless, as noted in the Anchorage Earthquake of 1964 

and again 30 years later in the Northridge Earthquake, soft soils do not necessarily 

affect the performance of the aboveground house structure as much as they affect the 

site and foundations (for example, by settlement, fissuring, or liquefaction).

TABLE 3.12 Site Soil Amplification Factor Relative to 

Acceleration

(short period, Site Class D) 

Short-Period Spectral 

Response Acceleration, 

SMS

≤ 0.25g 0.5g 0.75g 1.0g ≥ 1.25g

Site Soil Amplification 

Factor, Fa, 

1.6 1.4 1.2 1.1 1.0

The seismic response factor R has a long history in seismic design but with little in the way of 

scientific underpinnings until recently (FEMA, 2009b). In fact, the R factor can be traced 

back to expert opinion in the development of seismic design codes during the 1950s (ATC, 

1995). In recognition that buildings can effectively dissipate energy from seismic ground 

motions through ductile damage, the R factor was conceived to adjust the shear forces from 

that which would be experienced if a building could exhibit perfectly elastic behavior without 

some form of ductile energy dissipation (Chopra, 2012).

Those structural building systems that are able to withstand greater ductile damage and 

deformation without substantial loss of strength are assigned a higher value for R. The R 

factor also incorporates differences in dampening that are believed to occur for various 

structural systems. Table 3.13 provides some values for R that are relevant to residential 

construction. The Quantification of Building Seismic Performance Factors FEMA P-695

(FEMA, 2009b) methodology allows one to develop an R factor for a new LFRS based on the 

margin against collapse.
Residential Structural Design Guide 3-30

The overstrength factor Ω0 addresses the idea that a shear resisting system’s ultimate capacity 

usually is significantly higher than required by the design load as a result of intended safety 

margins. Designers incorporate overstrength factors in an attempt to address the principle of 

balanced design, striving to ensure that components such as connections have sufficient 

capacity to allow the LFRS to act in its intended ductile manner. These factors are applied at 

the load combination stage of force development.

The deflection amplification factor Cd is applied to adjust the deflection of story drift, which 

is determined by use of the seismic shear load as adjusted downward by the R factor. The use 

of this amplification factor will likely produce a conservative result of expected drift; drift 

calculations rarely are required in lowrise light-frame buildings because code-required drift 

limits have not been established for these structure types.

TABLE 3.13 Seismic Design Factors for Residential Construction

Structural System Response 

Modification 

Coefficient, R1

Overstrength 

factor, Ωo

Deflection 

Amplification 

Factor, Cd

Light-frame (wood) walls sheathed with wood 

structural panels rated for shear resistance or 

steel sheets

6.5 3 4

Light-frame shear walls with shear panels of 

all other materials

2.0 2.5 2

Special reinforced concrete shear walls2 5.0 2.5 5

Special reinforced masonry shear walls2 5.0 2.5 3.5

Ordinary plain concrete shear walls 1.5 2.5 1.5

Ordinary plain masonry shear walls 1.5 2.5 1.25

Notes:

1

The R factors may vary for a given structural system type depending on wall configuration, material selection, and connection detailing. 2

The wall is reinforced in accordance with concrete design requirements in ACI-318 or ACI-530. Nominally reinforced concrete or masonry that 

has conventional amounts of vertical reinforcement, such as one #5 rebar at openings and at 4 feet on center, may use the value for reinforced 

walls, provided the construction is no more than two stories above grade.

Design example 3.3 in section 3.12 demonstrates the calculation of design 

seismic shear load based on the simplified procedures (the reader is referred to 

chapter 6 for additional information on seismic loads and analysis).
Residential Structural Design Guide 3-31

3.8.3 Seismic Shear Force Distribution

As described in the previous section, the vertical distribution of seismic forces 

to separate stories on a light-frame building is assumed to be in accordance with the 

mass supported by each story. The lateral seismic force, Fx, induced at any level, is 

determined as

Equation 3.8-3

Fx = CvxV, and

Where:

Cvx = vertical distribution factor

V = total base shear

wi = portion of the total effective seismic weight of the structure at level i

wx = portion of the total effective seismic weight of the structure at level x

hi = height from the base to level i

hx = height from the base to level x

The horizontal distribution of seismic forces to various shear walls on a given 

story also varies in current practice for light-frame buildings. In chapter 6, several 

existing approaches to the design of the LFRS of light-frame houses address the issue 

of horizontal force distribution, with varying degrees of sophistication. Until methods 

of vertical and horizontal seismic force distribution are better understood for 

application to light-frame buildings, the importance of designer judgment cannot be 

overemphasized.

3.8.4 Other Seismic Design Considerations

Perhaps the single most important principle in seismic design is ensuring that 

the structural components and systems are adequately tied together to perform as a 

structural unit. Underlying this principle are a host of analytic challenges and 

uncertainties in actually defining what “adequately tied together” means in a 

repeatable, accurate, and theoretically sound manner.

Irregularities in the building shape are a key design consideration. 

Irregularites can occur in plan and in elevation. Plan irregularities can create torsional 

imbalances, thus requiring designs for moment distribution. Vertical irregularities are 

often stiffness irregularities, such as “soft stories,” “weak stories,” or “heavy stories.” 

Sometimes the vertical discontinuities can create unusual distribution of shear 

between LFRS.

When diaphragms are not flexible, the design should include the inherent torsional moment, 

Mt, resulting from the location of the structural masses plus the accidental torsional moment, 

Mta, caused by assumed displacement of the center of mass each way from its actual location 

by a distance equal to 5 percent of the dimension of the structure perpendicular to the 

direction of the applied forces. Overturning must be anticipated in the design, and the 

∑=

= n

i

i i

x x

vx

w h

w h C

1
Residential Structural Design Guide 3-32

structure must be designed to resist such forces. Story drift is computed as the largest 

difference of the deflections aligned vertically at the top and bottom of the story under 

consideration. The design story drift at level x is computed as















 =

e

d xe

x I

C δ δ

where Cd is the deflection amplification factor from table 3.12, xe δ is the deflection at the 

location of interest determined by elastic analysis, and Ie is the importance factor, which is 1.0 

for residential buildings.

For one- and two-story dwellings, the diaphragms are assumed to be flexible. Rigid 

diaphragms usually are those constructed of concrete or concrete-filled metal deck. ASCE 7-

10 has a set of conditions that are used to determine whether a diaphragm is flexible or rigid.

A key issue related to building damage involves deformation compatibility of materials and 

detailing in a constructed system. This issue may be handled through specification of 

materials that have similar deformation capabilities or by system detailing that improves 

compatibility. For example, a relatively flexible hold-down device installed near a rigid sill 

anchor causes greater stress concentration on the more rigid element, as evidenced by the 

splitting of wood sill plates in the Northridge Earthquake. The solution can involve increasing 

the rigidity of the hold-down device (which can lessen the ductility of the system, increase 

stiffness, and effectively increase seismic load) or redesigning the sill plate connection to 

accommodate the hold-down deformation and improve load distribution. Researchers in a 

FEMA-funded CUREE-CalTech project developed a solution for the sill plate connection: a 

3-inch-square washer for use on the sill plate anchor bolt. As a nonstructural example of 

deformation compatibility, gypsum board interior finishes crack in a major seismic event well 

before the structural capability of the wall’s structural sheathing is exhausted. Conversely, 

wood exterior siding and similar resilient finishes tend to deform compatibly with the wall 

and limit unacceptable visual damage (HUD, 1994).

3.9 Flood Loads

A significant level of construction occurs in the nation’s floodplains, so the design 

professional should be acquainted with the regulations and the design constraints required for 

building in these areas. The basic design premise for a flood condition is either to elevate the 

structure above the expected flood level or to build outside the regulatory floodplain. The full 

explanation of floodplain regulations is beyond the scope of this document; however, the 

following issues are important for all designers of buildings in floodplains:

• Floodplain regulations are local, so local zoning and/or building codes govern 

any construction in a floodplain.

• Local ordinances define the regulatory flood elevation, but it usually is the 

Base Flood Elevation (BFE), defined as the 1-percent annual exceedance 

probability flood. Frequently the community requires a certain number of feet 

of freeboard above the BFE to provide a margin of safety.
Residential Structural Design Guide 3-33

• The BFE usually is shown on a Flood Insurance Rate Map (FIRM), and this 

map is available either locally or digitally at FEMA’s Map Service Center.

• Minimum construction standards exist for buildings located in floodplains, 

and these standards govern elevation of the lowest living floor, the type of 

foundation that can be constructed, and the materials that can be used below 

the BFE.

Equations for flood loads are provided in ASCE 7, chapter 5, and for coastal flood conditions, 

flood formulas are available in FEMA’s Coastal Construction Manual (FEMA, 2011a) or in 

the United States Army Corps of Engineers Coastal Engineering Manual (USACE, 2009). 

Details specific to residential construction can be found in FEMA’s Home Builder’s Guide to 

Coastal Construction (FEMA, 2011b).

ASCE 24, the Flood Resistant Design and Construction standard, also contains significant 

flood design information. This standard does not include any information about flood loads 

but does suggest flood elevations for various building occupancies and provides design 

guidance for building issues from foundations to utilities.

3.10 Tornadoes

A tornado is a narrow, violently rotating column of air that extends from the 

base of a thunderstorm to the ground. They are the most violent of all atmospheric 

storms. Tornadoes occur in many parts of the world, but most of them by far occur in 

the United States, which experiences on average about 1,200 tornadoes per year 

(NCDC, 2013). Still, a direct tornado strike on a building is a relatively rare event, 

and the annual probability is lower than for other natural hazard events (that is 1.87E4 for a tornado strike of any intensity [Ramsdell et al., 2007] vs. 1.43E-3 for 

hurricane design wind speeds [ASCE, 2010]). Despite their small size, tornadoes can 

travel great distances and thus cause destruction to several communities within their 

path.

Building codes do not provide design guides for tornado loads for two 

reasons: (1) the rarity of the event and (2) the extreme magnitude of the tornado 

loads. The media report numerous opinions about whether building codes should 

cover that type of low probability, high consequence event. Many people believe that 

it would be economically unfeasible to design houses to resist the expected 200 mph 

and higher wind speeds produced in tornadoes. Substantial evidence also exists, 

however, that much damage could be reduced even when communities are struck by 

extremely violent (EF-4 and EF-5) tornadoes. The Enhanced Fujita (EF) scale is used 

to classify tornadoes by wind speed, using damage as the indicator of that speed. 

Examination of damage suggests that such extensive devastation is the result of 

inadequate structural systems in homes that were not designed for—and are incapable 

of resisting—any significant wind load. The damage report for a 1970 Lubbock 

tornado concluded that although the maximum estimated wind speed was 200 mph, 

the majority of building damage was caused by winds that were only in the 75-to-125 

mph range (Mehta et al., 1971).

The interest in developing tornado-resilient design of housing and other 

structures has gained interest recently following several years (for example, 2011 and 
Residential Structural Design Guide 3-34

2013) in which violent tornadoes have hit large, densely populated areas. The 

economic losses attributed to tornadoes since 2000 amount to 15 percent of the 

economic losses from hurricanes over that same period (NWS, 2013). Despite the low 

probability of tornado occurrences, the consequences are fairly high when a 

community is impacted by one of these natural hazard events.

The unique wind loads produced by an extreme tornado (that is, an EF-5 on 

the Enhanced Fujita scale) well exceed typical design wind loads, particularly in 

interior portions of the country, where tornadoes are most common but where design 

wind speeds typically are 115 mph. Most tornadoes, though, are not the most 

devastating kind; more than 90 percent of all tornadoes are classified as an EF-2 or 

lower on the EF Scale. Further, detailed analysis of the damage paths of recent violent 

tornadoes have shown that nearly 90 percent of the damage paths experience wind 

speeds at or less than the intensity of an EF-2 tornado (Prevatt et al., 2012). Applying 

the concepts used for hurricane design to buildings located in tornado-prone areas can 

reduce damage from the lowest wind speed tornadoes.

Tornado loads differ from typical straightline wind events such as hurricanes 

in that the loads are a superposition of the aerodynamic effects of the wind passing 

over and around the building and the significant pressure drop within the vortex of the 

tornado. In combination, these two effects can produce loads on the building in a 

tornado that are nearly three times higher than those for a straightline wind event with 

equivalent wind speed (Haan et al, 2010). Many factors affect the magnitude of 

tornado loads, however, including the tornado size, translation speed, approach angle, 

and air leakage through the impacted structure. The contributions of each effect have 

only recently been quantified. As a result, little current information exists for 

designing structures to survive tornado events. The next version of the commentary 

on the ASCE 7 standard will likely have some information that will help designers 

incorporate some level of tornado resistance in their designs. For the most severe 

events, such as those created by EF-4 or EF-5 tornadoes, a safe room or shelter built 

to FEMA guidance or ICC standards affords the best life safety protection. ICC 500 

(ICC, 2008) is a Standard for the Design and Construction of Storm Shelters; tornado 

safe room guidance is available in FEMA P-320 (FEMA, 2008b) and community 

shelter guidance is available in FEMA P-361 (FEMA, 2008a). These safe room and 

shelter guidance documents and standards have substantial design information about 

what wind speeds should be used for design and how to modify the wind pressure 

equation in ASCE 7 to accommodate the differences in the tornado wind structure 

compared to the hurricane or thunderstorm wind structure.

3.11 Other Load Conditions

In addition to the loads covered in sections 3.3 through 3.10 that are typically 

considered in the design of a home, other “forces of nature” may create loads on 

buildings. Some examples include

• frost heave;

• expansive soils; and

• temperature effects.
Residential Structural Design Guide 3-35

In certain cases, forces from these phenomena can drastically exceed 

reasonable design loads for homes. For example, frost heave forces can easily exceed 

10,000 psf (Linell and Lobacz, 1980). Similarly, the force of expanding clay soil can 

be impressive. The self-straining stresses induced by temperature-related expansion 

or contraction of a member or system that is restrained against movement can be very 

large, although those stresses are not typically a concern in wood-framed housing.

Sound design detailing is common practice to reduce or eliminate the load 

increases mentioned. For example, frost heave can be avoided by placing footings 

below a “safe” frost depth, building on nonfrost-susceptible materials, or using other 

frost protection methods (see chapter 4). Expansive soil loads can be avoided by 

isolating building foundations from expansive soil, supporting foundations on a 

system of deep pilings, and designing foundations that provide for differential ground 

movements. Temperature effects can be eliminated by providing construction joints 

that allow for expansion and contraction. Although such temperature effects on wood 

materials are practically negligible, some finishes, such as ceramic tile, can 

experience cracking when inadvertently restrained against small movements resulting 

from variations in temperature.

As noted at the beginning of this chapter, this guide does not address loads from ice, rain, and 

other exceptional sources. The reader is referred to ASCE 7 and other resources for 

information regarding special load conditions (ASCE, 2010).
Residential Structural Design Guide 3-36

3.12 Design Examples

EXAMPLE 3.1 Design Gravity Load Calculations and Use of ASD Load 

Combinations

Given Three-story conventional wood-framed home

28' x 44' plan, clear-span roof, floors supported at mid-span

Roof dead load = 15 psf (table 3.2) 

Wall dead load = 8 psf (table 3.2)

Floor dead load = 10 psf (table 3.2)

Roof snow load = 16 psf (section 3.7)

Attic live load = 10 psf (table 3.4)

Second- and third-floor live load = 30 psf (table 3.4)

First-floor live load = 40 psf (table 3.4)

Find Gravity load on first-story exterior bearing wall

Gravity load on a column supporting loads from two floors

Solution

1. Gravity load on first-story exterior bearing wall

• Determine loads on wall

Dead load = roof DL + 2 wall DL + 2 floor DL

= 1/2 (28 ft)(15 psf) + 2(8 ft)(8 psf) + 2(7 ft)(10 psf)

= 478 plf

Floor span assumes a center support wall, thus load on

exterior wall is 28'/4

Roof snow = 1/2(28 ft)(16 psf) = 224 plf

Live load = (30 psf + 30 psf)(7 ft) = 420 plf

(two floors)

Attic live load = (10 psf)(14 ft - 5 ft*) = 90 plf 

*edges of roof span not accessible to roof storage 

because of low clearance

• Apply applicable ASD load combinations (table 3.1)

D + 0.75L + 0.75S

Wall axial gravity load = 478 plf + 0.75*420 plf + 0.75*224 plf

= 961 plf*

*equals 1,029 plf if full attic live load allowance is 

included with L

The same load applies to the design of headers as well as to the wall studs. Of 

course, combined lateral (bending) and axial loads on the wall studs also must be 

checked (that is, D+W); refer to table 3.1 and example 3.2. For non-load-bearing 

exterior walls (that is, gable-end curtain walls), contributions from floor and roof 
Residential Structural Design Guide 3-37

live loads may be negligible (or significantly reduced), and the D+W load 

combination likely governs the design.

2. Gravity load on a column supporting a center floor girder carrying loads from two 

floors (first and second stories)

• Assume a column spacing of 16 ft

• Determine loads on column

(a) Dead load = Second floor + first floor + bearing wall supporting second 

floor

= (14 ft)(16 ft)(10 psf) + (14 ft)(16 ft)(10 psf) + (8 ft)(16 ft)(7 psf)

= 5,376 lbs

(b) Live load area reduction (equation 3.4-1)

- supported floor area = 2(14 ft)(16 ft) = 448 ft2 per floor

- reduction = 



 





+

4*448

15 0.25 = 0.6 > 0.5

OK

- first-floor live load = 0.6 (40 psf) = 24 psf

- second-floor live load = 0.6 (30 psf) = 18 psf

(c) Live load = (14 ft)(16 ft)[24 psf + 18 psf]

= 9,408 lbs

• Apply ASD load combinations (table 3.1)

The controlling load combination is D+L because the column supports no attic or 

roof loads. The total axial gravity design load on the column is 14,748 lbs (5,376 

lbs + 9,408 lbs).

Note: If LRFD material design specifications are used, the various loads would be 

factored in accordance with table 3.1; all other considerations and calculations 

remain unchanged.

EXAMPLE 3.2 Design Wind Load Calculations and Use of ASD Load 

Combinations

Given Site wind speed: 120 mph, gust 

Site wind exposure: suburban

Two-story home, 7:12 roof pitch, 28' x 44' plan (rectangular), gable roof, 12-inch 

overhang

Find Lateral (shear) load on lower-story end wall

Net roof uplift at connections to the side wall

Roof sheathing pull-off (suction) pressure

Wind load on a roof truss

Wind load on a rafter 
Residential Structural Design Guide 3-38

Lateral (out-of-plane) wind load on a wall stud

Solution

1. Lateral (shear) load on lower story end wall

Step 1: LRFD velocity pressure = 22 psf (table 3. 8)

Step 2: Adjusted velocity pressure (none required) = 22 psf

Step 3: Lateral roof coefficient = 0.8 (interpolated from table 3.9)

Lateral wall coefficient = 1.1 (table 3.9)

Step 4: Skip

Step 5: Determine design wind pressures

Roof projected area pressure = (22 psf)(0.8) = 17.6 psf (LRFD)

Wall projected area pressure = (22 psf)(1.1) = 24.2 psf (LRFD)

Now determine vertical projected areas (VPA) for lower-story end-wall tributary 

loading (assuming no contribution from interior walls in resisting lateral loads).

Roof VPA = [1/2 (building width)(roof pitch)] x [1/2 (building length)]

= [1/2 (28 ft)(7/12)] x [1/2 (44 ft)]

= [8.2 ft] x [22 ft]

= 180 ft2

Wall VPA = [(second-story wall height) + (thickness of floor) + 1/2 (firststory wall height)] x [1/2 (building length)]

= [8 ft + 1 ft + 4 ft] x [1/2 (44 ft)]

= [13 ft] x [22 ft]

= 286 ft2

Now determine shear load on the first-story end wall.

Shear = (roof VPA)(roof projected area pressure) + (wall VPA)(wall

projected area pressure)

= (180 ft2)(17.6 psf) + (286 ft2)(24.2 psf)

= 10,089 lbs (LRFD) or 10,089*0.6 = 6,053 lbs (ASD) 

The first-story end wall must be designed to transfer a shear load of 6,053 lbs. If 

side-wall loads were determined instead, the vertical projected area would include 

only the gable-end wall area and the triangular wall area formed by the roof. Use of 

a hip roof would reduce the shear load for the side and end walls.

2. Roof uplift at connection to the side wall (parallel-to-ridge)

Step 1: Velocity pressure = 22 psf (as before) (LRFD)

Step 2: Adjusted velocity pressure = 22 psf (as before)

Step 3: Skip

Step 4: Roof uplift pressure coefficient = -1.2 (table 3.10)

Roof overhang pressure coefficient = 0.7 (table 3.10)

Step 5: Determine design wind pressure

Roof horizontal projected area (HPA) pressure = -1.2 (22 psf)

= -24.2 psf

Roof overhang pressure = 0.7 (22 psf) = 15.4 psf (upward)

Now determine gross uplift at roof-wall reaction.
Residential Structural Design Guide 3-39

Gross uplift = 1/2 (roof span)(roof HPA pressure) + (overhang)(overhang 

pressure coefficient)

= 1/2 (30 ft)(-24.2 psf) + (1 ft)(-15.4 psf)

= -385 plf (upward)

Roof dead load reaction = 1/2 (roof span)(uniform dead load)

= 1/2 (30 ft)(15 psf*)

*table 3.2

= 225 plf (downward)

Now determine net design uplift load at roof-wall connection.

Net design uplift load = 0.6D + 0.6Wu (table 3.1)

= 0.6 (225 plf) + 0.6(-385 plf)

= -96 plf (net uplift)

The roof-wall connection must be capable of resisting a design uplift load of 96 plf. 

Generally, a toenail connection will meet the design requirement, depending on the 

nail type, nail size, number of nails, and density of wall framing lumber (see 

chapter 7). At high design wind speeds or in more open wind exposure conditions, 

roof tie-down straps, brackets, or other connectors should be considered.

3. Roof sheathing pull-off (suction) pressure

Step 1: Velocity pressure = 22 psf (as before)

Step 2: Adjusted velocity pressure = 22 psf (as before)

Step 3: Skip

Step 4: Roof sheathing pressure coefficient (suction) = -2.8 (table 3.10)

Step 5: Roof sheathing pressure (suction) = (22 psf)(-2.8)

= -61.6 psf

The fastener load depends on the spacing of roof framing and spacing of the 

fastener. Fasteners in the interior of the roof sheathing panel usually have the 

largest tributary area and therefore are critical. Assuming 24-inch-on-center roof 

framing, the fastener withdrawal load for a 12-inch-on-center fastener spacing is as 

follows:

Fastener withdrawal load = (fastener spacing)(framing spacing)

(roof sheathing pressure)

= (1 ft)(2 ft)(-61.6 psf)

= -123.2 lbs (LRFD) or 0.6*123.2 

= 73.9 lbs (ASD)

At high wind conditions, a closer fastener spacing or higher capacity fastener (that

is, deformed shank nail) may be required; refer to chapter 7.

4. Load on a roof truss

Step 1: Velocity pressure = 22 psf (as before)

Step 2: Adjusted velocity pressure = 22 psf (as before)

Step 3: Skip

Step 4: Roof truss pressure coefficient = -0.9, +0.4 (table 3.10)

Step 5: Determine design wind pressures

(a) Uplift = -0.9 (22 psf) = -19.8 psf
Residential Structural Design Guide 3-40

(b) Inward = 0.4 (22 psf) = 8.8 psf

Because the inward wind pressure is less than the minimum roof live load (that is, 

15 psf, table 3.4), the following load combinations would govern the roof truss 

design, and and the D+W load combination could be dismissed (refer to table 3.1):

D + (Lr or S)

0.6D + 0.6Wu*

*The net uplift load for truss design is relatively small in this 

case (approximately 4.9 psf).

5. Load on a rafter

Step 1: Velocity pressure = 22 psf (as before)

Step 2: Adjusted velocity pressure = 22 psf (as before)

Step 3: Skip

Step 4: Rafter pressure coefficient = -1.2, +0.7 (table 3.10)

Step 5: Determine design wind pressures

(a) Uplift = (-1.2)(22 psf) = -26.4 psf

(b) Inward = (0.7)(22 psf) = 15.4 psf

Rafters in cathedral ceilings are sloped, simply supported beams, whereas rafters 

that are framed with cross-ties (that is, ceiling joists) constitute a component (that 

is, top chord) of a site-built truss system. Assuming the former in this case, the 

rafter should be designed as a sloped beam by using the span measured along the 

slope. By inspection, the minimum roof live load (D+Lr) governs the design of the 

rafter in comparison to the wind load combinations (see table 3.1). The load 

combination 0.6 D+0.6Wu can be dismissed in this case for rafter sizing but must 

be considered when investigating wind uplift for the rafter-to-wall and rafter-toridge beam connections.

6. Lateral (out-of-plane) wind load on a wall stud

Step 1: Velocity pressure = 22 psf (as before)

Step 2: Adjusted velocity pressure = 22 psf (as before)

Step 3: Skip

Step 4: Wall stud pressure coefficient = -1.2, +1.1 (table 3.10)

Step 5: Determine design wind pressures

(a) Outward = (-1.5)(22 psf) = -33.0 psf

(b) Inward = (1.1)(22 psf) = 24.2 psf

Obviously, the outward pressure of 33.0 psf governs the out-of-plane bending load 

design of the wall stud. Because the load is a lateral pressure (not uplift), the 

applicable load combination is D+W (refer to table 3.1), resulting in a combined 

axial and bending load. The axial load would include the tributary building dead 

load from supported assemblies (that is, walls, floors, and roof). The bending load 

would then be determined by using the wind pressure of 33.0 psf applied to the 

stud as a uniform line load on a simply supported beam, calculated as follows:

Uniform line load, w = (wind pressure)(stud spacing)

= (33.0 psf)(1.33 ft*)

*assumes stud spacing of 16 inches on center

= 43.9 plf (LRFD)
Residential Structural Design Guide 3-41

Of course, the following gravity load combination would also need to be 

considered in the stud design (refer to table 3.1):

D + 0.75 L + 0.75 (Lr or S)

The stud is actually part of a wall system (that is, sheathing and interior finish) and 

can add substantially to the calculated bending capacity; refer to chapter 5.

EXAMPLE 3.3 Design Earthquake Load Calculation

Given Site ground motion, Ss = 1g

Site soil condition = firm (default)

Roof snow load < 30 psf

Two-story home, 28' x 44' plan, typical construction

Find Design seismic shear on first-story end wall, assuming no interior shear walls or 

contribution from partition walls

Solution

1. Determine tributary mass (weight) of building to first-story seismic shear.

Roof dead load = (28 ft)(44 ft)(15 psf) = 18,480 lb

Second-story exterior wall dead load = (144 lf)(8 ft)(8 psf) = 9,216 lb

Second-story partition wall dead load = (28 ft)(44 ft)(6 psf) = 7,392 lb

Second-story floor dead load = (28 ft)(44 ft)(10 psf) = 12,320 lb

First-story exterior walls (1/2 height) = (144 lf)(4 ft)(8 psf) = 4,608 lb

Assume first-story interior partition walls are capable of supporting at least the 

seismic shear produced by their own weight

Total tributary weight = 52,016 lb

2. Determine total seismic story shear on first story.

SDS = 2/3 (Ss)(Fa) (equation 3.8-2)

= 2/3 (1.0g)(1.1) (Fa = 1.1 from table 3.12)

= 0.74 g

V = W (equation 3.8-1)

= 5.5

1.2 (0.74g)

(52,016 lb) (R = 5.5 from table 3.13)

= 8,399 lb

3. Determine design shear load on the 28-foot end walls.

Assume that the building mass is evenly distributed and that stiffness is also 

reasonably balanced between the two end walls; refer to chapter 6 for additional 

guidance.

With the above assumption, the load is simply distributed to the end walls 
Residential Structural Design Guide 3-42

according to tributary weight (or plan area) of the building; therefore,

End wall shear = 1/2 (8,399 lb) = 4,200 lb

Note that the design shear load from wind (100 mph gust, exposure B) in example 

3.2 is somewhat greater (5,912 lbs).

3.13 References

American Concrete Institute (ACI). 2011. Building Code Requirements for Structural 

Concrete and Commentary, ACI Standard 318. Farmington Hills, MI: ACI.

American Society of Civil Engineers (ASCE). 2010. Minimum Design Loads for 

Buildings and Other Structures, ASCE Standard 7-10. Reston, VA: ASCE.

American Society of Testing and Materials (ASTM). 2005. Standard Test Method for 

Performance of Exterior Windows, Curtain Walls, Doors, and Impact 

Protective Systems Impacted by Missile(s) and Exposed to Cyclic Pressure 

Differentials (ASTM E1886-05). West Conshohocken, PA: ASTM.

———. 2009a. Standard Specification for Asphalt-Saturated Organic Felt Used in 

Roofing and Waterproofing, Type II (ASTM D226-09). West Conshohocken, 

PA: ASTM.

———. 2009B. STANDARD SPECIFICATION FOR PERFORMANCE OF 

EXTERIOR WINDOWS, CURTAIN WALLS, DOORS, AND IMPACT 

PROTECTIVE SYSTEMS IMPACTED BY WINDBORNE DEBRIS IN 

HURRICANES (ASTM E1996). WEST CONSHOHOCKEN, PA: ASTM.

———. 2009c. Standard Specification for Self-Adhering Polymer Modified 

Bituminous Sheet Materials Used as Steep Roof Underlayment for Ice Dam 

Protection (ASTM D1970-09) West Conshohocken, PA: ASTM.

———. 2011. Standard Test Method for Wind Resistance of Asphalt Shingles

(ASTM D7158-11). West Conshohocken, PA: ASTM.

———. 2012. Standard Test Method for Laboratory Compaction Characteristics of 

Soil Using Standard Effort (D 698-12). West Conshohocken, PA: ASTM.

Applied Technology Council (ATC). 1995. A Critical Review of Current Approaches 

to Earthquake-Resistant Design (ATC-34). Redwood City, CA: ATC.

American Wood Council, 2012. Wood Frame Construction Manual for One- and 

Two-Family Dwellings (ANSI/AWC WFCM 2012) Leesburg, VA: AWC.
Residential Structural Design Guide 3-43

Chalk, Philip L., and Ross B. Corotis. 1978. “Probability Models for Design Live 

Loads,” Journal of the Structural Division 106 (ST10).

Chopra, Anil K. 2012. Dynamics of Structures, Theory and Applications to 

Earthquake Engineering, 4th ed. Upper Saddle River, NJ: Prentice-Hall.

Ellingwood, B., J.G. MacGregor, T.V. Galambos, and C.A. Cornell. 1982. 

“Probability Based Load Criteria: Load Factors and Load Combinations,” 

Journal of the Structural Division 108 (ST5). Reston, VA: American Society 

of Civil Engineers.

Federal Emergency Management Agency (FEMA). 2008a. Design and Construction 

Guidance for Community Safe Rooms, FEMA P-361. Washington, DC: 

FEMA.

———. 2008b. Taking Shelter From the Storm: Building a Safe Room For Your 

Home or Small Business, FEMA P-320. Washington, DC: FEMA.

———. 2009a. National Earthquake Hazards Reduction Program (NEHRP) 

Recommended Seismic Provisions for New Buildings and Other Structures.

Washington, DC: FEMA.

———. 2009b. Quantification of Building Seismic Performance Factors, FEMA P695. Washington, DC: FEMA.

———. 2011a. Coastal Construction Manual FEMA P-55. Washington, DC: FEMA.

———. 2011b. Home Builder’s Guide to Coastal Construction, FEMA P-499.

Washington, DC: FEMA.

Fernandez, G., F.J. Masters, and K.R. Gurley. 2010. “Performance of Hurricane 

Shutters Under Impact by Roof Tiles,” Engineering Structures 32: 3384–

3393.

Filiatrault et al. 2010. “Experimental Seismic Response of a Full-Scale Light-Frame 

Wood Building,” Journal of Structural Engineering 136 (3): 246-254.

Galambos, T.V. and B. Ellingwood. 1986. “Serviceability Limit States: Deflection,”

Journal of Structural Engineering 112 (1): 67–84.

Galambos, T.V., B. Ellingwood, J.G. MacGregor, and C.A. Cornell. 1982. 

“Probability Based Load Criteria: Assessment of Current Design Practice,” 

Journal of the Structural Division 108 (5): 959–977.
Residential Structural Design Guide 3-44

Haan et al. 2010. “Tornado-Induced Wind Loads on a Low-Rise Building,” Journal 

of Structural Engineering 136 (1): 106–116.

Harris, Michael E., Ross B. Corotis, and Carl J. Bova. 1981. “Area-Dependent 

Processes for Structural Live Loads,” Journal of the Structural Division 107 

(5): 857–872.

U.S. Department of Housing and Urban Development (HUD). 1977. Manual for the 

Construction of Residential Basements in Non-Coastal Flood Environments. 

Prepared for HUD by the NAHB Research Foundation, Rockville, MD.

———. 1994. Assessment of Damage to Residential Buildings Caused by the 

Northridge Earthquake. Prepared for HUD by the NAHB Research Center, 

Inc., Upper Marlboro, MD.

———. 2011. Safer, Stronger Homes: Reducing Your Risk During a Natural 

Disaster. Prepared for HUD by Newport Partners LLC, Davidsonville, MD,

and ARES Consulting, West River, MD.

International Code Council, Inc. (ICC). 2008. ICC/NSSA Standard for the Design and 

Construction of Storm Shelters, ICC 500. Washington, DC: ICC.

———. 2011. 2012 International Residential Code for One- and Two-Family 

Dwellings. Country Club Hills, IL: ICC.

———. 2012. 2012 International Building Code. Washington, DC: ICC.

Linell, K.A., and E.F. Lobacz. 1980. Design and Construction of Foundations in 

Areas of Deep Seasonal Frost and Permafrost, Special Report 80-34.

Hanover, NH: Cold Regions Research and Engineering Laboratory.

Masters, F.J., K.R. Gurley, N. Shah*, and G. Fernandez*. 2010. “Vulnerability of 

Residential Window Glass to Lightweight Windborne Debris.” Engineering 

Structures 32 (4): 911–921.

Mehta, K.C., J.R. McDonald, J.E. Minor, and A.J. Sanger. 1971. “Response of 

Structural Systems to the Lubbock Storm; Final Report,” storm research 

report, Texas Tech University, Lubbock.

Merritt, Frederick S. 1983. Standard Handbook for Civil Engineers, 3rd ed. New 

York: McGraw-Hill.

National Climatic Data Center (NCDC). 2013. Average annual number of tornadoes. 

Available at https://www.ncdc.noaa.gov/climate-information/extremeevents/us-tornado-climatology
Residential Structural Design Guide 3-45

National Weather Service (NWS). 2013. Natural Hazard Statistics (1995-2012),

compiled by the National Climatic Data Center. Available at 

http://www.nws.noaa.gov/os/hazstats.shtml (accessed November 5).

Peck, R.B., W.E. Hanson, and T.H. Thornburn. 1974. Foundation Engineering. New 

York: John Wiley & Sons. 

Prevatt et al. 2012. Making the case for improved structural design: Tornado 

outbreaks of 2011. Leadership and Management in Engineering 12 (4), 254-

270.

Ramsdell, J.V., J.P. Rishel, U.S. Nuclear Regulatory Commission, and Pacific 

Northwest Laboratory. 2007. “Tornado Climatology of the Contiguous United 

States.” Washington, DC: U.S. Nuclear Regulatory Commission, Office of 

Nuclear Regulatory Research, Division of Risk Assessment and Special 

Projects.

Rankine, W., 1857. On the Stability of Loose Earth. Proceedings of the Royal Society 

of London, 147:9-27.

Salzano, C.T., F.J. Masters, and J.D. Katsaros, 2010. Water Penetration Resistance of 

Residential Window Installation Options for Hurricane-Prone Areas. Building 

and Environment 45 (6): 1373–1388.

University of Alberta. 1992. Thin Wall Foundation Testing. Department of Civil 

Engineering, University of Alberta.

United States Army Corps of Engineers (USACE). 2009. Coastal Engineering 

Manual. Vicksburg, MS: USACE.
Residential Structural Design Guide 4-1

CHAPTER 4

Design of Foundations

4.1 General

A foundation supports and anchors the superstructure of a building and 

transfers all loads (including those from flood, wind, or seismic events) imposed on it 

directly to the ground. Foundations distribute the loads to the earth over an adequate 

area so that loads do not exceed the bearing capacity of the soil and so that lateral 

movement or settlement is minimized. In cold climates, the bottom of the foundation 

must be below the frost line to prevent freeze-thaw damage and frost heave of the 

footing.

A foundation in residential construction may consist of a footing, wall, slab, 

pier, pile, or a combination of these elements. This chapter addresses the following 

foundation types—

• Crawl space.

• Basement.

• Slab-on-grade with stem wall.

• Monolithic slab.

• Piles.

• Piers.

• Alternative methods.

As discussed in chapter 1, the most common residential foundation materials 

are cast-in-place concrete and concrete masonry (that is, concrete block). 

Preservative-treated wood, precast concrete, and other materials may also be used. 

The concrete slab-on-grade is a prevalent foundation type in the South and 

Southwest; basements are the most common type in the East and Midwest. Crawl 

spaces are common in the Northwest and Southeast. Pile foundations designed to 

function after being exposed to scour and erosion are commonly used in coastal flood 

zones to elevate structures above flood levels. Piles also are used in weak or 
Residential Structural Design Guide 4-2

expansive soils to reach a stable stratum and on steeply sloped sites. Figure 4.1 

depicts different foundation types; a brief description follows.

A crawl space is a building foundation that uses a perimeter foundation wall 

to create an underfloor space that is not habitable; the interior crawl space elevation 

may or may not be below the exterior finish grade. In mapped flood plains, a crawl 

space that has the interior grade below the exterior grade on all sides is considered a 

basement. A basement typically is defined as a portion of a building that is partly or 

completely below the exterior grade and that may be used as habitable space, as 

storage space, or for parking. The primary difference between a basement and a crawl 

space is height (basements usually are taller). The floors of basements usually are 

finished, and the interiors frequently are finished. The wall height is sometimes 

determined by the depth of the footing required for frost protection. 

A slab-on-grade with an independent stem wall is a concrete floor supported 

by the soil independently of the rest of the building. The stem wall supports the 

building loads and in turn is supported directly by the soil or a footing. A monolithic 

or thickened-edge slab is a ground-supported slab-on-grade with an integral footing 

(that is, a thickened edge); it is normally used in warmer regions that have little or no 

frost depth but is also used in colder climates when adequate frost protection is 

provided (see section 4.7). 

When necessary, piles are used to transmit the load to a deeper soil stratum 

with a higher bearing capacity to prevent failure from undercutting of the foundation 

by scour from flood water flow at high velocities and to elevate the building above 

required flood elevations. Piles also are used to isolate the structure from expansive 

soil movements. 

Pier and beam foundations can provide an economical alternative to crawl 

space perimeter wall construction. A common practice is to use a brick curtain wall 

between piers for appearance and bracing.

The design procedures and information in this chapter cover the following 

topics.

• Foundation materials and properties.

• Soil-bearing capacity and footing size.

• Concrete or gravel footings.

• Concrete and masonry foundation walls.

• Preservative-treated wood walls.

• Insulating concrete foundations.

• Concrete slabs on grade.

• Pile foundations.

• Frost protection.

Concrete design procedures generally follow the strength design method 

contained in the American Concrete Institute’s ACI 318 (ACI, 2011) although certain 

aspects of the procedures may be considered conservative relative to conventional 

residential foundation applications. For this reason, this guide provides supplemental 

design guidance when practical and technically justified. ACI 332 (ACI, 2010), which 

contains design provisions and guidance specific to residential construction, is 
Residential Structural Design Guide 4-3

referenced in the International Residential Code (IRC) as an alternative to the 

conventional foundation requirements of the code or the design procedures of ACI 

318. Masonry design procedures follow the allowable stress design (ASD) method of 

The Masonry Society’s TMS 402 (TMS, 2011). Wood design procedures are used to 

design the connections between the foundation system and the structure above and 

follow the ASD method for wood construction (see chapter 7 for connection design 

information). In addition, the designer is referred to the applicable design standards 

for symbol definitions and additional guidance because the intent of this chapter is to 

provide supplemental instruction in the efficient design of residential foundations.

To maintain consistency, this guide uses the load and resistance factor design 

(LRFD) load combinations that were used in chapter 3, which are also those specified 

in the American Society of Civil Engineers’ ASCE 7. There may be some minor 

variations in those required in ACI 318 for strength design of concrete. The purpose 

of this guide is to provide designs that are at least consistent with current residential 

building code and construction practice. With respect to the design of concrete in 

residential foundations, the guide seeks to provide reasonable safety margins that 

meet or exceed the minimums required for other, more crucial requirements of a 

home—namely, the safety of lives. The designer is responsible for ensuring that the 

design meets the building code requirements and will be approved by the building 

official.
Residential Structural Design Guide 4-4

FIGURE 4.1 Types of Foundations
Residential Structural Design Guide 4-5

4.2 Material Properties

A residential designer using concrete and masonry materials must have a basic 

understanding of such materials, including variations in the materials’ composition 

and structural properties. In addition, a designer must take into consideration soils, 

which are also considered a foundation material (Section 4.3 provides information on 

soil bearing). A brief discussion of the properties of concrete and masonry follows.

4.2.1 Concrete

The concrete compressive strength (fc') used in residential construction is 

typically either 2,500 or 3,000 pounds per square inch (psi), although other values 

may be specified. For example, 4,000 psi concrete may be used for improved 

weathering resistance in particularly severe climates or unusual applications. The 

concrete compressive strength may be verified in accordance with the American 

Society for Testing and Materials’ ASTM C39 (ASTM, 2012c). Given that the rate of 

increase in concrete strength diminishes with time, the specified compressive strength 

usually is associated with the strength attained after 28 days of curing time, when the 

concrete attains about 85 percent of its fully cured compressive strength.

Concrete is a mixture of cement, water, and sand, gravel, crushed rock, or 

other aggregates. Sometimes one or more admixtures are added to change certain 

characteristics of the concrete, such as workability, durability, and time of hardening. 

The proportions of the components determine the concrete mix’s compressive 

strength and durability.

Type

Portland cement is classified into several types, in accordance with ASTM 

C150 (ASTM, 2012b). Residential foundation walls typically are constructed with 

Type I cement, which is a general-purpose Portland cement used for the vast majority 

of construction projects. Other types of cement are appropriate in accommodating 

conditions related to heat of hydration in massive pours and sulfate resistance. In 

some regions, sulfates in soils have caused durability problems with concrete. The 

designer should check into local conditions and practices.

Weight

The weight of concrete varies depending on the type of aggregates used in the 

concrete mix. Concrete typically is classified as lightweight or normal weight. The 

density of unreinforced normal weight concrete ranges between 144 and 156 pounds 

per cubic foot (pcf) and typically is assumed to be 150 pcf. Residential foundations 

usually are constructed with normal weight concrete.
Residential Structural Design Guide 4-6

Slump

Slump is the measure of concrete consistency; the higher the slump, the wetter 

the concrete and the easier it flows. Slump is measured in accordance with ASTM 

C143 (ASTM, 2012d) by inverting a standard 12-inch-high metal cone, filling it with 

concrete, and then removing the cone; the amount the concrete that settles in units of 

inches is the slump. Most foundations, slabs, and walls consolidated by hand methods 

have a slump between 4 and 6 inches. One problem associated with a high-slump 

concrete is segregation of the aggregate, which leads to cracking and scaling. 

Therefore, a slump of greater than 6 inches should be avoided. Adding water lowers 

the strength while improving workability, so the total amount of water in the concrete 

should be carefully monitored and controlled. Admixtures used during extremely cold 

or hot weather placement (or for other reasons) may change the slump.

Weather Resistance

Concrete is largely weather resistant. When concrete may be subjected to 

freezing and thawing during construction, however, or when concrete is located in 

regions prone to extended periods of freezing, additional measures must be taken. 

Those requirements can be found in the IRC (ICC, 2012), and include air entrainment 

and increased minimum compressive strength requirements. 

Admixtures

Admixtures are materials added to the concrete mix to improve workability 

and durability and to retard or accelerate curing. Some of the most common 

admixtures are described below.

• Water reducers improve the workability of concrete without reducing its 

strength.

• Retarders are used in hot weather to allow more time for placing and 

finishing concrete. Retarders may also reduce the early strength of 

concrete.

• Accelerators reduce the setting time, allowing less time for placing and 

finishing concrete. Accelerators may also increase the early strength of 

concrete.

• Air entrainers are used for concrete that will be exposed to freeze-thaw 

conditions and deicing salts. Less water is needed and segregation of 

aggregate is reduced when air entrainers are added.
Residential Structural Design Guide 4-7

Reinforcement

Concrete has high compressive strength but low tensile strength; therefore, 

reinforcing steel often is embedded in the concrete to provide additional tensile 

strength and ductility. In the rare event that the capacity is exceeded, the reinforcing 

steel begins to yield, thereby preventing an abrupt failure that may otherwise occur 

with plain, unreinforced concrete. For this reason, a larger safety margin is used in the 

design of plain concrete construction than in reinforced concrete construction.

Steel reinforcement is available in grade 40 or grade 60; the grade number 

refers to the minimum tensile yield strength (fy) of the steel (i.e., grade 40 is a 

minimum 40 thousand pounds per square inch [ksi] steel and grade 60 is a minimum 

60 ksi steel). Either grade may be used for residential construction; however, most 

steel reinforcement in the U.S. market today is grade 60. The concrete mix, or slump, 

must be adjusted by adding the appropriate amount of water to allow the concrete to 

flow easily around the reinforcement bars, particularly when the bars are closely 

spaced or are crowded at points of overlap. Close rebar spacing rarely is required in 

residential construction, however, and should be avoided in design if at all possible.

The most common steel reinforcement or rebar sizes in residential 

construction are No. 3, No. 4, and No. 5, which correspond to diameters of 3/8 inch, 

1/2 inch, and 5/8 inch, respectively. The bar designations indicate the bar size in 1/8-

inch increments. These three sizes of rebar are easily manipulated at the jobsite by 

using manual bending and cutting devices. Table 4.1 shows useful relationships 

between the rebar number, diameter, and cross-sectional area for reinforced concrete 

and masonry design.

Fiber reinforcement is being used in some concrete slab installation. The fiber 

could be steel, natural, or synthetic. It helps (1) improve resistance to freeze-thaw, (2) 

increase resistance to some spalling of the surface, (3) control cracking, and (4) 

improve the concrete’s shatter resistance. Fibers generally do not increase the 

structural strength of the concrete slab and do not replace normal reinforcing bars 

used for tensile strength.

TABLE 4.1 Rebar Size, Diameter, and Cross-Sectional Areas

Size Diameter (inches) Area (square inches)

No. 3 3/8 0.11

No. 4 1/2 0.20

No. 5 5/8 0.31

No. 6 3/4 0.44

No. 7 7/8 0.60

No. 8 1 0.79
Residential Structural Design Guide 4-8

4.2.2 Concrete Masonry Units

Concrete masonry units (CMUs), commonly referred to as concrete blocks, are 

composed of Portland cement, aggregate, and water. In some situations, CMUs may 

also include admixtures. Low-slump concrete is molded and cured to produce strong 

blocks or units. Residential foundation walls typically are constructed with units 7 5/8 

inches (nominal 8 inches) high by 15 5/8 inches (nominal 16 inches) long, providing 

a 3/8-inch allowance for the width of mortar joints. Nominal 8- and 12-inch-thick 

CMUs are readily available for use in residential construction. 

Type

ASTM C90 (ASTM, 2013) requires that the minimum average design strength 

(f'm) of standard CMUs be 1,900 psi, with no individual unit having a compressive 

strength of less than 1,700 psi. Higher strengths also may be specified if required by 

design. The ASTM classification includes two types. Type II is a non-moisturecontrolled unit and is the type typically used for residential foundation walls.

Weight

CMUs are available with different densities by altering the type(s) of 

aggregate used in their manufacture. CMUs typically are referred to as lightweight, 

medium weight, or normal weight, with respective unit weights or densities of less 

than 105 pcf, between 105 and 125 pcf, and more than 125 pcf. Residential 

foundation walls typically are constructed with low- to medium-weight units because 

of the low compressive strength required. Lower density units are generally more 

porous, however, and must be properly protected to resist moisture intrusion. A 

common practice in residential basement foundation wall construction is to provide a 

cement-based parge coating and a brush- or spray-applied bituminous coating on the 

belowground portions of the wall. Section R406 in the IRC provides the prescriptive 

requirements for parging and damp-proofing or waterproofing foundation walls that 

retain earth and enclose interior spaces. The parge coating is not required for concrete 

foundation wall construction.

Hollow or Solid

CMUs are classified as hollow or solid in accordance with ASTM C90. The 

net concrete cross-sectional area of most CMUs ranges from 50 to 70 percent, 

depending on unit width, face-shell and web thicknesses, and core configuration. 

Hollow units are defined as those in which the net concrete cross-sectional area is less 

than 75 percent of the gross cross-sectional area. Solid units are not necessarily solid 

but are defined as those in which the net concrete cross-sectional area is 75 percent of 

the gross cross-sectional area or greater.
Residential Structural Design Guide 4-9

Mortar

Masonry mortar is used to join CMUs into a structural wall; it also retards air 

and moisture infiltration. The most common way to lay block is in a running bond 

pattern, in which the vertical head joints between blocks are offset by half the block 

length from one course to the next. Mortar is composed of water, cement, lime, and 

clean, well-graded sand, and water and is typically classified into types M, S, N, O, 

and K, in accordance with ASTM C270 (ASTM, 2012a). Residential foundation walls 

typically are constructed with type M or type S mortar, both of which are generally 

recommended for load-bearing interior and exterior walls, including above- and 

below-grade applications.

Grout

Grout is a slurry consisting of cementitious material, aggregate, and water. 

When needed, grout commonly is placed in the hollow cores of CMUs to provide a 

wall with added strength. In reinforced load-bearing masonry wall construction, grout 

is usually placed only in those hollow cores containing steel reinforcement. The grout 

bonds the masonry units and steel so that they act as a composite unit to resist 

imposed loads. Grout may also be used in unreinforced concrete masonry walls for 

added strength. The IRC requires grouted cells at foundation sill and sole plate anchor 

bolt locations, regardless of whether the masonry wall is otherwise reinforced.

4.3 Soil-Bearing Capacity

and Footing Size

Soil-bearing investigations rarely are required for residential construction 

except when a history of local problems provides evidence of known risks (for 

example, organic deposits, landfills, expansive soils, and seismic risk). Soil-bearing 

tests on stronger-than-average soils can, however, justify using smaller footings or 

eliminating footings entirely if the foundation wall provides sufficient bearing 

surface. Table 4.2 provides a conservative relationship between soil type and loadbearing value. A similar table is published in the building codes (table R401.4.1 in the 

IRC). These presumptive soil-bearing values, however, should be used only when the 

building codes do not require geotechnical investigation reports (section R401.4, 

IRC). 
Residential Structural Design Guide 4-10

TABLE 4.2 Presumptive Soil-Bearing Values by Soil Description

Presumptive Load-Bearing 

Value (psf) Soil Description

1,500 Clay, sandy clay, silty clay, clayey silt, silt, and sandy silt

2,000 Sand, silty sand, clayey sand, silty gravel, and clayey gravel

3,000 Gravel and sandy gravel

4,000 Sedimentary and foliated rock

12,000 Crystalline bedrock

psf = pounds per square foot.

Source: ICC (2012).

When a soil-bearing investigation is desired to determine more accurate and 

economical footing requirements, the designer commonly turns to ASTM D1586, 

Standard Penetration Test (SPT) and Split-Barrel Sampling of Soils (ASTM, 2011). 

This test relies on a 2-inch-diameter device driven into the ground with a 140-pound 

hammer dropped from a distance of 30 inches. The number of hammer drops or 

blows needed to create a 1-foot penetration (blow count) yields values that can be 

roughly correlated to soil-bearing values, as shown in Table 4.3. The instrumentation 

and cost of conducting the SPT usually are not warranted for typical residential 

applications. Nonetheless, the SPT method provides information on deeper soil strata 

and thus can offer valuable guidance for foundation design and building location, 

particularly when subsurface conditions threaten to be problematic. The values in 

Table 4.3 are associated with the blow count from the SPT method. Many engineers 

can provide reasonable estimates of soil bearing by using smaller penetrometers at 

lower cost, although such devices and methods may require an independent 

calibration to determine presumptive soil-bearing values and may not be able to 

detect deep subsurface problems. Calibrations may be provided by the manufacturer 

or, alternatively, developed by the engineer.

In addition to ASTM D1586, the Dynamic Cone Penetrometer (DCP) test 

(Burnham and Johnson, 1993), has gained widespread use as a more economical 

alternative with equivalent accuracy. In this handheld test, a metal cone is driven into 

the ground by repeatedly striking it with a 17.6-pound (8-kilogram) weight, dropped 

from a distance of 2.26 feet (575 millimeters). Penetration of the cone is measured 

after each blow; the blow count per 1 3/4-inch penetration is approximately 

equivalent to the SPT blow count provided in table 4.3.

The designer should exercise judgment when selecting the final design value 

and be prepared to make adjustments (increases or decreases) in interpreting and 

applying the results to a specific design. The values in tables 4.2 and 4.3 generally are 

associated with a safety factor of 3 (Naval Facilities Engineering Command, 1986) 

and are considered appropriate for noncontinuous or independent spread footings 

supporting columns or piers (that is, point loads). Use of a safety factor could be 
Residential Structural Design Guide 4-11

considered for smaller structures with continuous spread footings, such as houses, or 

structures for which ultimate (LRFD) values are used for design loads. The 

presumptive values in Table 4.3—or as modified as described previously—are 

intended to be used with the ASD load combinations in chapter 3. If LRFD (strength) 

design load combinations are used, then the presumptive values should be 

additionally adjusted (that is, divided by the maximum load factor in the load 

combination considered, usually a factor of 1.6 for live or snow loads).

Table 4.3 Presumptive Soil-Bearing Values (psf) Based on Standard 

Penetrometer Blow Count

In Situ Consistency, N1

Loose2

(5 to 10 blows per 

foot)

Firm

(10 to 25 blows per 

foot)

Compact

(25 to 50 blows per 

foot)

Noncohesive Soils

Gravel 4,000 (10) 8,000 (25) 11,000 (50)

Sand 2,500 (6) 5,000 (20) 6,000 (35)

Fine sand 1,000 (5) 3,000 (12) 5,000 (30)

Silt 500 (5) 2,000 (15) 4,000 (35)

In Situ Consistency, N1:

Soft3

(3 to 5 blows per 

foot)

Medium

(about 10 blows 

per foot)

Stiff

(more than 20 

blows per foot)

Cohesive 

Soils

Clay, sand, gravel mixtures 2,000 (3) 5,000 (10) 8,000 (20)

Sandy or silty clay 1,000 (4) 3,000 (8) 6,000 (20)

Clay 500 (5) 2,000 (10) 4,000 (25)

Source: Naval Facilities Engineering Command, 1986.

psf = pounds per square foot.

1

N denotes the standard penetrometer blow count in blows per foot, in accordance with ASTM D1586; shown in parentheses.

2

Compaction should be considered in these conditions, particularly when the blow count is five blows per foot or less. 3

Pile and grade beam foundations should be considered in these conditions, particularly when the blow count is five blows per foot or less.

The required width or area of a spread footing is determined by dividing the 

building load on the footing by the soil-bearing capacity from table 4.2 or table 4.3, as 

shown below. Building design loads, including dead and live loads, should be 

determined in accordance with chapter 3 by using ASD load combinations.
Residential Structural Design Guide 4-12

Soil bearing capacity in psf

Load in lbs Areaindependent spread footing − =

Soil bearing capacity in psf

Load lb per linear foot plf Widthcontinuous footing − = , ( )

4.4 Footings

The objectives of footing design are—

• To provide a level surface for construction of the foundation wall.

• To provide adequate transfer and distribution of building loads to the 

underlying soil.

• To provide adequate strength, in addition to the foundation wall, to 

prevent differential settlement of the building in weak or uncertain soil 

conditions by bridging those poor soil conditions.

• To place the building foundation at a sufficient depth to avoid frost 

heave or thaw weakening in frost-susceptible soils and to avoid 

organic surface soil layers.

• To provide adequate anchorage or mass (when needed in addition to 

the foundation wall) to resist potential uplift, sliding, and overturning 

forces resulting from high winds or severe seismic events.

This section presents design methods for concrete and gravel footings. The 

designer must first establish the required footing width in accordance with section 

4.3. Further, if soil conditions are stable or the foundation wall can adequately 

resist potential differential settlement, the footing may be completely eliminated.

By far, the most common footing in residential construction is a 

continuous concrete spread footing; however, concrete and gravel footings are 

both recognized in prescriptive footing size tables in residential building codes for 

most typical conditions (ICC, 2012). In contrast, special conditions give rise to 

engineering concerns that must be addressed to ensure the adequacy of any 

foundation design. Special conditions include—

• Steeply sloped sites requiring a stepped footing.

• High wind conditions.

• Inland or coastal flooding conditions.

• High-hazard seismic conditions.

• Poor soil conditions.
Residential Structural Design Guide 4-13

4.4.1 Simple Gravel and Concrete Footing Design

Building codes for residential construction contain tables that prescribe 

minimum footing widths for plain concrete footings (ICC, 2012). Alternatively, 

footing widths may be determined in accordance with section 4.3, based on a 

site’s particular loading condition and presumptive soil-bearing capacity. The 

following are general rules of thumb for determining the thickness of plain 

concrete footings for residential structures, once the required bearing width has 

been calculated.

• The minimum footing thickness should not be less than the distance 

the footing extends outward from the edge of the foundation wall or 6 

inches, whichever is greater.

• The footing width should project a minimum of 2 inches from both 

faces of the wall (to allow for a minimum construction tolerance) but 

not greater than the footing thickness.

These rules of thumb generally result in a footing design that differs 

somewhat from the plain concrete design provisions of chapter 22 of ACI 318.

Footing widths generally follow the width increments of standard excavation 

equipment (in other words, a backhoe bucket size of 12, 16, or 24 inches). 

Although longitudinal steel reinforcement is not always required for residentialscale structures in typical soil conditions, the designer should consider adding 

some (two No. 4 or No. 5 bars is common) to avoid possible footing cracking 

where soil consolidation or a loss of soil-bearing capacity can occur. For 

situations in which the rules of thumb or prescriptive code tables do not apply or 

in which a more economical solution is possible, a more detailed footing analysis 

may be considered (see section 4.4.2). Example 4.1 in section 4.9 illustrates a 

plain concrete footing design in accordance with the simple method described 

herein.

Much like a concrete footing, a gravel footing may be used to distribute 

foundation loads to a sufficient soil-bearing surface area. A gravel footing 

provides a continuous path for water or moisture and thus must be drained in 

accordance with the foundation drainage provisions of the IRC. Gravel footings 

are constructed of crushed stone or gravel that is consolidated by tamping or 

vibrating. Pea gravel, which is naturally consolidated, does not require 

compaction and can be screeded to a smooth, level surface, much like concrete. 

Although typically associated with pressure-treated wood foundations (refer to 

section 4.5.3), a gravel footing can support cast-in-place or precast concrete 

foundation walls.

The size of a gravel footing usually is based on a 30- to 45-degree angle of 

repose for distributing loads; therefore, as with plain concrete footings, the 

required depth and width of the gravel footing depends on the width of the 

foundation wall, the foundation load, and soil-bearing values. Following a rule of 

thumb similar to that for a concrete footing, the gravel footing thickness should be 
Residential Structural Design Guide 4-14

no less than 1.5 times its extension beyond the edge of the foundation wall or, in 

the case of a pressure-treated wood foundation, the mud sill. Just as with a 

concrete footing, the thickness of a gravel footing may be considered in meeting 

the required frost depth. In soils that are not naturally well drained, provision 

should be made to adequately drain a gravel footing.

4.4.2 Concrete Footing Design

For many residential footing designs, prescriptive and conventional 

residential footing requirements found in residential building codes and 

construction guides are adequate, if not conservative. Concrete design for 

residential construction is covered in ACI 332 Residential Code Requirements for 

Structural Concrete (ACI, 2010). To improve performance and economy or to 

address peculiar conditions, however, a footing may need to be specially 

designed. Many floor plans in today’s residential buildings are partially open and 

frequently create nonuniform loading conditions on load-bearing walls and 

footings. These nonuniform load conditions must be considered in the design of 

footings, and reliance on strictly prescriptive methods of design is not always a 

sound design decision.

A footing is designed to resist the upward-acting pressure created by the 

soil beneath the footing; that pressure tends to make the footing bend upward at 

its edges. According to ACI 318, the three modes of failure considered in 

reinforced concrete footing design are one-way shear, two-way shear, and flexure 

(see figure 4.2). Bearing (crushing) is also a possible failure mode but is rarely 

applicable to residential loading conditions. To simplify calculations for the three 

failure modes, the following discussion explains the relation of the failure modes 

to the design of plain and reinforced concrete footings (Refer to ACI 318 for 

additional commentary and guidance). The design equations used later in this 

section are based on ACI 318 and principles of engineering mechanics, as 

described herein. Moreover, the approach is based on the assumption of uniform 

soil-bearing pressure on the bottom of the footing; therefore, walls and columns 

should be supported as close as possible to the center of the footings.

One-Way (Beam) Shear

When a footing fails due to one-way (beam) shear, the failure occurs at an 

angle approximately 45 degrees to the wall, as shown in figure 4.2. For plain 

concrete footings, the soil-bearing pressure has a negligible effect on the diagonal 

shear tension for distance t from the wall edge toward the footing edge; for 

reinforced concrete footings, the distance used is d, which equals the depth to the 

footing rebar (see figure 4.2). As a result, one-way shear is checked by assuming 

that beam action occurs at a critical failure plane extending across the footing 

width, as shown in figure 4.2. One-way shear must be considered in similar 

fashion in both continuous wall and rectangular footings; however, for ease of 

calculation, continuous wall footing design typically is based on one lineal foot of 

wall or footing. 
Residential Structural Design Guide 4-15

Two-Way (Punching) Shear

When a footing fails by two-way (punching) shear, the failure occurs at an 

angle approximately 30 degrees to the column or pier, as shown in figure 4.2. 

Punching shear rarely is a concern in the design of continuous wall footings; thus, 

punching shear is usually checked only in the case of rectangular or circular 

footings with a heavily loaded pier or column that creates a large concentrated 

load on a relatively small area of the footing. For plain concrete footings, the soilbearing pressure has a negligible effect on the diagonal shear tension at distance 

t/2 from the face of a column toward the footing edges; for reinforced concrete 

footings, the distance from the face of the column is d/2 (see figure 4.2). The 

shear force, therefore, consists of the net upward-acting pressure on the area of 

the footing outside the punched-out area (hatched area in figure 4.2). For square, 

circular, or rectangular footings, shear is checked at the critical section that 

extends in a plane around a concrete, masonry, wood, or steel column or pier that 

forms the perimeter bo of the area previously described. 
Residential Structural Design Guide 4-16

FIGURE 4.2 Critical Failure Planes in Continuous or Square Concrete

Spread Footings
Residential Structural Design Guide 4-17

Flexure (Bending)

The maximum moment in a footing deformed by the upward-acting soil 

pressures would logically occur in the middle of the footing; however, the rigidity 

of the wall or column above resists some of the upward-acting forces and affects 

the location of maximum moment. As a result, the critical flexure plane for 

footings supporting a rigid wall or column is assumed to be located at the face of 

the wall or column. Flexure in a concrete footing is checked by computing the 

moment created by the soil-bearing forces acting over the cantilevered area of the 

footing that extends from the critical flexure plane to the edge of the footing 

(hatched area in figure 4.2). The approach for masonry walls in ACI 318 differs 

slightly in that the failure plane is assumed to be located one-fourth of the way 

under a masonry wall or column, creating a slightly longer cantilever. For the 

purpose of this guide, the difference is considered unnecessary. 

Bearing Strength

Conditions in which concrete bearing or compressive strength is a concern 

are uncommon in typical residential construction; therefore, a design check 

usually can be dismissed as “OK by inspection.” In rare and peculiar instances in 

which bearing compressive forces on the concrete are extreme and approach or 

exceed the specified concrete compressive strength, the designer should consult 

ACI 318 for appropriate guidance.

4.4.2.1 Plain Concrete Footing Design

In this section, the design of plain concrete footings is presented by using 

the concepts related to shear and bending covered in the previous section (refer to 

example 4.1 in section 4.9 for a design example of a plain concrete footing).

Shear

In the equations that follow for one- and two-way shear, the dimensions 

are in accordance with figure 4.2; units of inches should be used. ACI 318 

requires that the overall thickness (t) be taken as 2 inches less than the actual 

thickness to compensate for uneven trench conditions. The following equations 

are specifically tailored for footings supporting walls or square columns because 

such footings are common in residential construction. The equations may be 

generalized for use with other conditions (for example, rectangular footings and 

rectangular columns, round footings) by following the same principles. In 

addition, the terms 4/3 c f' and 4 c f' are in units of psi and represent lower 

bound estimates of the ultimate shear stress capacity of unreinforced concrete.
Residential Structural Design Guide 4-18

ACI 318•22.5,22.7

One-Way (Beam) Shear

φ Vc ≥ Vu basic design check for shear

V (q )(0.5(b T) t) u = s − − factored shear load (lb)

b

P

q u s =

f lt

3

4 Vc c φ = φ ′ factored shear capacity (lb)

φ = 0.65 resistance factor

Two-Way (Punching) Shear

φ Vc ≥ Vu basic design check for shear

( )( ( ) ) 2

u s V = q b − T + t shear load (lb) due to factored load Pu (lb)

b

P

q u s =

V 4 f b t c c o φ = φ ′ factored shear capacity (lb) 

b 4(T t) o = +

φ = 0.65 resistance factor

Flexure

For a plain concrete footing, flexure (bending) is checked by using the 

equations that follow for footings that support walls or square columns (see figure 

4.2). The dimensions in the equations are in accordance with figure 4.2 and use 

units of inches. The term c 5 f' is in psi and represents a lower bound estimate of 

the ultimate tensile (rupture) stress of unreinforced concrete in bending.

uniform soil bearing pressure (psi) due to 

factored foundation load Pu (lb)

uniform soil bearing pressure (psi) due to 

factored foundation load Pu (lb)

perimeter of critical failure plane 

around a square column or pier
Residential Structural Design Guide 4-19

ACI 318•22.5,22.7

φMn ≥ Mu basic design check for bending

( )

2

u qs b T

8

1 M =  −

bl

P

q u s =

φMn = φ5 f c

′S factored moment capacity (in-lb) for plain concrete

2

t

6

1 S =  section modulus (in3

) for footing

φ = 0.65 resistance factor for plain concrete in bending

4.4.2.2 Reinforced Concrete Footing Design

For situations in residential construction in which a plain concrete footing 

may not be practical or in which reducing the footing thickness is more 

economical, steel reinforcement should be considered. A reinforced concrete 

footing is designed similar to a plain concrete footing; however, the concrete 

depth d to the reinforcing bar is used to check shear instead of the entire footing 

thickness t. The depth of the rebar is equal to the thickness of the footing minus 

the diameter of the rebar db and the concrete cover c. In addition, the moment 

capacity is determined differently due to the presence of the reinforcement, which 

resists the tension stresses induced by the bending moment. Finally, a higher 

resistance factor reflects the more consistent bending strength of reinforced 

concrete compared to unreinforced concrete. 

As specified by ACI 318, a minimum of 3 inches of concrete cover over 

steel reinforcement is required when concrete is in contact with soil. In addition, 

ACI 318 does not permit a depth d less than 6 inches for reinforced footings 

supported by soil. The designer may relax these limits, provided that the strength 

analysis demonstrates adequate capacity; however, a reinforced footing thickness 

of significantly less than 6 inches may be considered impractical, even though it 

may calculate acceptably. One exception may be found where a nominal 4-inchthick slab is reinforced to serve as an integral footing for an interior load-bearing 

wall (which is not intended to transmit uplift forces from a shear wall overturning 

restraint anchorage in high-hazard wind or seismic regions). Further, the concrete 

cover should not be less than 2 inches for residential applications, although this 

recommendation may be somewhat conservative for interior footings that are 

generally less exposed to ground moisture and other corrosive agents. Example 

4.2 of section 4.9 illustrates reinforced concrete footing design. The placement of 

steel to comply with concrete cover requirements may also significantly reduce 

the depth of steel, thus reducing flexural capacity of the concrete element; the 

designer must consider this reduced depth of steel. 

factored moment (in-lb) due to soil pressure qs (psi) acting on 

cantilevered portion of footing

uniform soil bearing pressure (psi) due to factored load Pu

(lb)
Residential Structural Design Guide 4-20

Shear

In the following equations for one- and two-way shear, the dimensions are 

in accordance with figure 4.2; units of inches should be used. Shear reinforcement 

(that is, stirrups) is usually considered impractical for residential footing 

construction; therefore, the concrete is designed to withstand the shear stress, as 

expressed in the equations. The equations are specifically tailored for footings 

supporting walls or square columns because such footings are common in 

residential construction. The equations may be generalized for use with other 

conditions (rectangular footings and rectangular columns, round footings, and so 

on) by following the same principles. In addition, the terms 2 c f' and 4 c f' are 

in units of psi and represent lower bound estimates of the ultimate shear stress 

capacity of reinforced concrete. 

ACI 318•11.12,15.5 

One-Way (Beam) Shear

φ Vc ≥ Vu basic design check for shear

V (q )(0.5(b T) d) u = s − −

b

P

q u s =

V 2 f d c c φ = φ ′  factored shear capacity (lb)

5d b d = t − c − 0. depth of reinforcement

φ = 0.85

Two-Way (Punching) Shear

φ Vc ≥ Vu basic design check for shear

( ( ) ) u 2

u b T d

b

P

V − + 



 





 = 

 shear load (lb) due to factored load Pu (lb)

V 4 f b d c c o φ = φ ′ factored shear capacity (lb)

b 4(T d) o = +

φ = 0.85

resistance factor for reinforced concrete in 

shear

perimeter of punching shear failure plane 

around a square column or pier

resistance factor for reinforced concrete in 

shear

shear load (lb) due to uniform soil-bearing 

pressure, qs (psi)

uniform solid-bearing pressure (psi) due to 

factored foundation load Pu (lb)
Residential Structural Design Guide 4-21

Flexure

The flexure equations that follow pertain specifically to reinforced 

concrete footings that support walls or square columns. The equations may be 

generalized for use with other conditions (rectangular footings and rectangular 

columns, round footings, and so on) by following the same principles. The 

alternative equation for nominal moment strength Mn is derived from force and 

moment equilibrium principles by using the provisions of ACI 318. Most 

designers are familiar with the alternative equation that uses the reinforcement 

ratio ρ and the nominal strength coefficient of resistance Rn. The coefficient is 

derived from the design check that ensures that the factored moment (due to 

factored loads) Mu is less than the factored nominal moment strength φ Mn of the 

reinforced concrete. To aid the designer in short-cutting these calculations, design 

manuals provide design tables that correlate the nominal strength coefficient of 

resistance Rn to the reinforcement ratio ρ for a specific concrete compressive 

strength and steel yield strength.

ACI 318•15.4

φ Mn ≥ Mu basic design check for bending

2

u s q (b T) 8

1 M =  −

factored moment (in-lb) due to soil pressure 

q(psi) acting on cantilevered portion of the 

footing s

)

2

a M A f (d φ n = φ s y − factored nominal moment capacity (in-lb) 

0.85 f' c

A f

a s y =

(l is substituted for the ACI 318 symbol b for 

the concrete beam width and is consistent with 

the footing dimensioning in figure 4.2)

φ = 0.9 resistance factor for reinforced concrete in 

bending

Alternate method to determine Mn

















′

ρ φ =φρ −

c

y

n y 0.85f

0.5 d f

M bdf d

















′ − 













 ′ ρ =

c

n

y

c

0.85f

2R

f

0.85f

 reinforcement ratio determined by use of Rn 

nominal strength “coefficient of resistance

2

u n d

M R

φ =

l is substituted for the ACI 318 symbol b for 

the concrete beam width and is consistent with 

the footing dimensioning in figure 4.2)

As = ρ d

defines reinforcement ratio ρ

(l is substituted for the ACI 318 symbol b for 

the concrete beam width and is consistent with 

the footing dimensioning in figure 4.2)
Residential Structural Design Guide 4-22

Minimum Reinforcement

Because of concerns with shrinkage and temperature cracking, ACI 318 

requires a minimum amount of steel reinforcement. For grade 60 reinforcing steel, 

the minimum area of steel used for shrinkage and temperature cracking is 0.0018 

square inches. ACI 318 requirements on the minimum area of steel for flexural 

members are shown in the following equations:

ACI 318•7.12, 10.5

y min f ρ = 200 or 0.0018 or 0.0018

As,min = ρmin ld (l is substituted for the ACI 318 symbol b for the concrete beam 

width and is consistent with the footing dimensioning in figure 

4.2)

Designers often specify one or two longitudinal No. 4 or No. 5 bars for 

wall footings as nominal reinforcement when building on questionable soils, 

when required to maintain continuity of stepped footings on sloped sites, or when 

conditions result in a changed footing depth. For most residential foundations, 

however, the primary resistance against differential settlement is provided by the 

deep beam action of the foundation wall, especially if the wall is reinforced 

masonry or concrete; footing reinforcement may provide limited benefit. In such 

cases, the footing simply acts as a platform for the wall construction and 

distributes loads to a larger soil-bearing area.

Lap Splices

Where reinforcement cannot be installed in one length to meet 

reinforcement requirements, as in continuous wall footings, reinforcement bars 

must be lapped to develop the bars’ full tensile capacity across the splice. In 

accordance with ACI 318, a minimum lap length of 40 times the diameter of the 

reinforcement bar is required for splices in the reinforcement. In addition, the 

separation between spliced or lapped bars must not exceed eight times the 

diameter of the reinforcement bar or 6 inches, whichever is less. This is a design 

or construction issue that frequently causes failures during extreme loading 

conditions from high winds, storm surge, or seismic events. In accordance with 

TMS 402, the maximum distance between lapped or spliced bars is one-fifth the 

splice length or 8 inches, whichever is less.

For foundation systems consisting of a plain concrete footing and a plain 

concrete stem wall, a minimum of one bar should be provided at the top of the 

stem wall and at the bottom of the footing. Plain concrete footings supporting 

walls are permitted in Seismic Design categories A, B or C without longitudinal 

reinforcement. For buildings located in Seismic Categories D or E, the footings 

should have at least two continuous longitudinal reinforcing bars no smaller than 

No. 4 and must have a total area of not less than 0.002 times the gross crosssectional area of the footing. Footings more than 8 inches (203 millimeters) thick 
Residential Structural Design Guide 4-23

must have a minimum of one bar at the top and bottom of the footing. Corners 

and intersections must have continuity of reinforcement.

4.5 Foundation Walls

The objectives of foundation wall design are—

• To transfer the load of the building to the footing or directly to the 

earth.

• To provide adequate strength, in combination with the footing 

when required, to prevent differential settlement.

• To provide adequate resistance to shear and bending stresses 

resulting from lateral soil pressure.

• To provide anchorage for the above-grade structure to resist wind 

or seismic forces.

• To provide a moisture-resistant barrier to below-ground habitable 

space, in accordance with the building code.

• To isolate non-moisture-resistant building materials from the 

ground.

In some cases, masonry or concrete foundation walls incorporate a 

nominal amount of steel reinforcement to control cracking. Engineering 

specifications generally require reinforcement of concrete or masonry foundation 

walls because of somewhat arbitrary limits on minimum steel-to-concrete ratios, 

even for plain concrete walls. Residential foundation walls are generally 

constructed of unreinforced or nominally reinforced concrete or masonry or of 

preservative-treated wood, however. The nominal reinforcement approach has 

provided many serviceable structures. This section addresses the issue of 

reinforcement and presents rational design approaches for residential concrete and 

masonry foundation walls. 

In most cases, a designer may select a design for concrete or concrete 

masonry walls from the prescriptive tables in the applicable residential building 

code or the IRC (ICC, 2012). Sometimes, however, a specific design applied with 

reasonable engineering judgment results in a more efficient and economical 

solution than that prescribed by the codes. The designer may elect to design the 

wall as either a reinforced or plain concrete wall. The following sections detail 

design methods for both wall types.

4.5.1 Concrete Foundation Walls

Regardless of the type of concrete foundation wall selected, the designer 

must determine the nominal and factored loads that, in turn, govern the type of 

wall (that is, reinforced or unreinforced) that may be appropriate for a given 

application. The following LRFD load combinations suggested for the design of 

residential concrete foundation walls are based on table 3.1 of chapter 3:
Residential Structural Design Guide 4-24

• 1.4 D + 1.6 H

• 1.2 D + 1.6 H + 1.6 L + 0.5 (Lr or S)

• 1.2 D + 1.6 H + 1.6 (Lr or S) + 0.5 L

In light-frame homes, the first load combination typically governs 

foundation wall design. Axial load increases moment capacity of concrete walls 

when they are not appreciably eccentric, as is the case in typical residential 

construction.

To simplify the calculations, the designer may conservatively assume that 

the foundation wall acts as a simple span beam with pinned ends, although such 

an assumption will tend to over predict the stresses in the wall. In any event, the 

simple span model requires that the wall be adequately supported at its top by the 

connection to the floor framing and at its base by the connection to the footing or 

bearing against a basement floor slab. Appendix A contains basic load diagrams 

and beam equations to assist the designer in analyzing typical loading conditions 

and element-based structural actions encountered in residential design. Once the 

loads are known, the designer can perform design checks for various stresses by 

following ACI 318 and the recommendations contained herein.

As a practical consideration, residential designers must keep in mind that 

concrete foundation walls typically are a nominal 6, 8, or 10 inches thick. The 

typical concrete compressive strength used in residential construction is 2,500 or 

3,000 psi, although other strengths are available. Table 4.4 illustrates 

recommended minimum concrete compressive strengths based on use and 

weathering potential. Typical reinforcement tensile yield strength is 60,000 psi 

(grade 60) and is primarily a matter of market supply (Refer to section 4.2.1 for 

more information on concrete and steel reinforcement material properties).

Table 4.4 Minimum Compressive Strength f'c at 28 Days and 

Maximum Slump of Concrete

Type or location of concrete 

construction Weathering Probability Maximum 

Negligible slump, in. (mm)

f'c, psi (MPa)

Moderate

f'c, psi (MPa)

Severe

f'c, psi (MPa)

Type 1: Walls and foundations not 

exposed to weather; interior slabs-ongrade, not including garage floor slabs2500 (17) 2500 (17) 2,500 (17) 6(150)

Type 2: Walls, foundations, and other 

concrete work exposed to weather, 

except as noted below

2500 (17) 3000 (21) 3000 (21) 6(150)

Type 3: Driveways, curbs, walkways, 

patios, porches, steps, and stairs 

exposed to weather; garage floors, 

slabs

2500 (17) 3500 (24) 4500 (31) 5(125)

f'c = minimum compressive strength

mm = millimeters. 

MPa = megapascal

psi = pounds per square inch
Residential Structural Design Guide 4-25

4.5.1.1 Plain Concrete Wall Design

ACI 318 defines “plain concrete” as structural concrete with no 

reinforcement or with less reinforcement than the minimum amount specified for 

reinforced concrete, and ACI 318•22.0 permits its use in wall design. Structural 

plain concrete basement, foundation, or other walls below the base are permitted 

in detached one- and two-family, stud-bearing wall dwellings three stories or 

fewer in height. Plain concrete walls must be used only in regions of low to 

moderate seismic risk—Seismic Design Category A, B, or C; otherwise, 

reinforcing is required. ACI 318 recommends incorporating contraction and 

isolation joints to control cracking; however, this is not a typical practice for 

residential foundation walls. Temperature and shrinkage cracking is practically 

unavoidable but is considered to have negligible impact on the structural integrity 

of a residential wall. Cracking can be controlled (that is, minimizing potential 

crack widening) by reasonable use of horizontal reinforcement; chapter 4 of the 

IRC (ICC, 2012) provides some specific prescriptive requirements governing 

reinforcement size and spacing in plain concrete foundation walls.

ACI 318 limits plain concrete wall thickness to a minimum of 7.5 inches; 

however, the IRC (ICC, 2012) permits nominal 6-inch-thick foundation walls 

when the height of unbalanced fill is less than a prescribed maximum.

Adequate strength must be provided and should be demonstrated by 

analysis, in accordance with the ACI 318 design equations and the 

recommendations of this section. Depending on soil loads, analysis should 

confirm conventional residential foundation wall practice in typical conditions 

(Refer to example 4.3 of section 4.9 for an illustration of a plain concrete 

foundation wall design).

Shear Capacity

Shear stress is a result of the lateral loads on a structure associated with 

wind, earthquake, or backfill forces. Lateral loads are either normal to the wall 

surface (that is, perpendicular or out of plane) or parallel to the wall surface (that 

is, in plane). The designer must consider both perpendicular and parallel shear in 

the wall.

Perpendicular shear is rarely a controlling factor in the design of 

residential concrete foundation walls except for some foundation walls with 

substantial backfill loads. Parallel shear also is usually not a controlling factor in 

residential foundation walls except for walls that are shear walls resisting lateral 

loads from high winds or seismic events.

If greater shear capacity is required in a plain concrete wall, increasing the 

wall thickness or increasing the concrete compressive strength may accomplish 

that purpose. Alternatively, a wall can be reinforced in accordance with section 

4.5.1.2.

The following equations apply to both perpendicular and parallel shear, in 

conjunction with figure 4.3, for plain concrete walls. For parallel shear, the 

equations do not address overturning and bending action that occurs in a direction 
Residential Structural Design Guide 4-26

parallel to the wall, particularly for short segments of walls under significant 

parallel shear load. For concrete foundation walls, that is generally not a concern; 

for above-grade wood-frame walls, the concern is addressed in chapter 6 in detail.

ACI 318•22.5.4

Vu ≤ φVn

Vu = maximum factored shear load on the wall

f bh

3

4 Vn c φ = φ ′

φ = 0.65

FIGURE 4.3 Shear Calculations for

Plain Concrete Walls: Variables Defined
Residential Structural Design Guide 4-27

Combined Axial Load and Bending Capacity

The following ACI 318 equations account for the combined effects of 

axial load and bending moment on a plain concrete wall. The intent is to ensure 

that the concrete face in compression and the concrete face in tension resulting 

from factored nominal axial and bending loads do not exceed the factored 

nominal capacity for concrete. Example 4.4 of section 4.9 demonstrates a method 

of plotting the interaction equation that follows. (Refer to section 4.5.1.3 for 

information on interaction diagrams.)

ACI 318•22.5.3, 22.6.3

1

M

M

P

P

n

u

n

u ≤

φ

+

φ

on the compression face

5 f ' A

P

S

M c

g

u u − ≤ φ on the tension face

Mu > Mu,min

Mu = maximum factored nominal moment on wall

u,min 1hPu M = 0.

Mn = 0.85f'

c S

g

2

c n c A

32h

l

P 0.6f' 1 





















 





 = −

Pu = factored nominal axial load on the wall at point of maximum moment

φ = 0.65

Even though a plain concrete wall often calculates as adequate, the 

designer may elect to add a nominal amount of reinforcement for crack control or 

other reasons. Walls determined inadequate to withstand combined axial load and 

bending moment may gain greater capacity through increased wall thickness or 

increased concrete compressive strength. Alternatively, the wall may be 

reinforced in accordance with section 4.5.1.2. Walls determined to have adequate 

strength to withstand shear and combined axial load and bending moment may 

also be checked for deflection, but this is usually not a limiting factor for typical 

residential foundation walls.

4.5.1.2 Reinforced Concrete Design

ACI 318 allows two approaches to the design of reinforced concrete, with 

some limits on wall thickness and the minimum amount of steel reinforcement; 

however, ACI 318 also permits these requirements to be waived in the event that 

structural analysis demonstrates adequate strength and stability in accordance 

with ACI 318•14.2.7 (refer to examples 4.4 in section 4.9 for the design of a 

reinforced concrete foundation wall).

Reinforced concrete walls should be designed in accordance with ACI 

318•14.4 by using the strength design method. The following checks for shear 
Residential Structural Design Guide 4-28

and combined flexure and axial load determine if a wall is adequate to resist the 

applied loads.

Shear Capacity

Shear stress is a result of the lateral loads on a structure associated with 

wind, earthquake, or lateral soil forces. The loads are either normal to the wall 

surface (that is, perpendicular or out of plane), however, or parallel to the wall 

surface (that is, in plane). The designer must check both perpendicular and 

parallel shear in the wall to determine if the wall can resist the lateral loads 

present.

If greater shear capacity is required, it may be obtained by (1) increasing 

the wall thickness, (2) increasing the concrete compressive strength, (3) adding 

horizontal shear reinforcement, or (4) installing vertical reinforcement to resist 

shear through shear friction. Shear friction is the transfer of shear through friction 

between two faces of a crack. Shear friction also relies on resistance from 

protruding portions of concrete on either side of the crack and by dowel action of 

the reinforcement that crosses the crack. The maximum limit on reinforcement 

spacing of 12 or 24 inches specified in ACI 318•11.5.4 is considered to be an 

arbitrary limit. When reinforcement is required, practical experience dictates 48 

inches as an adequate maximum spacing for residential foundation wall design.

The following equations provide checks for both perpendicular and 

parallel shear in conjunction with figure 4.4. For parallel shear, the equations do 

not address overturning and bending action that occurs in a direction parallel to 

the wall, particularly for short segments of walls under significant parallel shear 

load. For concrete foundation walls, that generally is not a concern; for abovegrade wood-framed walls, the topic is addressed in chapter 6 in detail.

ACI 318•11.5,11.7, 11.10

Vu ≤ φVn

Vu = maximum factored shear load on wall

Vn = Vc + Vs

V 2 f b d c c w = ′

8 f b d s

A f d

V c w

v y

s = ≤ ′ when Vu > φ Vc when 

φ = 0.85

Shear-Friction Method

Vu ≤ φVn

n vf y cAc V = A f μ ≤ 0.2f ′ and Ac ≤ 800 and 

Ac = bwh

φ = 0.85
Residential Structural Design Guide 4-29

FIGURE 4.4 Shear Calculations

in Reinforced Concrete Walls: Variables Defined
Residential Structural Design Guide 4-30

Combined Flexural and Axial Load Capacity

ACI 318 prescribes reinforcement requirements for concrete walls. 

Foundation walls commonly resist both an applied axial load from the structure 

above and an applied lateral soil load from backfill. To ensure that the wall’s 

strength is sufficient, the designer must first determine slenderness effects (that is, 

Euler buckling) in the wall. ACI 318•10.10 provides an approximation method to 

account for slenderness effects in the wall; however, the slenderness ratio must 

not be greater than 100. The slenderness ratio is defined in the following section 

as the ratio between unsupported length and the radius of gyration. In residential 

construction, the approximation method, more commonly known as the moment 

magnifier method, is usually adequate because slenderness ratios typically are less 

than 100 in foundation walls.

The moment magnifier method uses the relationship of the axial load and 

lateral load in addition to wall thickness and unbraced height to determine a 

multiplier of 1 or greater, which accounts for slenderness in the wall. The 

multiplier is termed the moment magnifier. It magnifies the calculated moment in 

the wall resulting from the lateral soil load and any eccentricity in axial load. 

Together, the axial load and magnified moment are used to determine whether the 

foundation wall section is adequate to resist the applied loads. The following steps 

are required to determine the amount of reinforcement required in a typical 

residential concrete foundation wall to resist combined flexure and axial loads—

• Calculate axial and lateral loads.

• Verify that the nonsway condition applies.

• Calculate slenderness.

• Calculate the moment magnifier.

• Plot the axial load and magnified moment on an interaction diagram.

The following sections discuss the procedure in detail.

Slenderness

Conservatively, assuming that the wall is pinned at the top and bottom, 

slenderness in the wall can be calculated by using the equation that follows. The 

effective length factor k is conservatively assumed to equal 1 in this condition. A 

value of k less than 1 (for example, 0.7) may actually better represent the end 

conditions (that is, nonpinned state) of residential foundation walls.
Residential Structural Design Guide 4-31

ACI 318•10.10

34

r

klu < slenderness ratio

12

d

bd

12

bd

A

I r

2 3

= = = radius of gyration

Moment Magnifier Method

The moment magnifier method is an approximation method allowed in ACI 

318•10.10 for concrete walls with a slenderness ratio less than or equal to 100. If the 

slenderness ratio is less than 34, then the moment magnifier is equal to 1 and 

requires no additional analysis. The design procedure and equations that follow align 

with ACI 318•10.12. The equation for EI, as listed in ACI 318, is applicable to walls 

containing a double layer of steel reinforcement. Residential walls typically contain 

only one layer of steel reinforcement; therefore, the equation for EI, as listed herein, 

is based on section 10.12 (ACI, 2008).

ACI 318•10.12.3

Mu,mag = δMu ⇐ Magnified Moment

1

0.75P

P 1

C

c

u

m ≥





 



 −

δ =

( )

2

u

2

c

kl

EI P π =

Cm = 0.6

or

Cm = 1 for members with transverse loads between supports

M P (0.6 0.03h) u,min = u +

( )

β

≥

β

− ≥

β = c g c g c g 0.1E I h

e 0.4E I E I 0.5

EI

u

2

P

M e =

0.9 0.5 12 1.0 2 β = + βd − ρ ≥

g

s

A

A

ρ =

u

u,dead

d P

P

β =

c

1.5

c c c E = 57,000 f ′ or w 33 f ′

Given that the total factored axial load in residential construction typically 

falls below 3,000 pounds per linear foot of wall and that concrete compressive 

strength typically is 3,000 psi, table 4.5 provides prescriptive moment magnifiers. 
Residential Structural Design Guide 4-32

Interpolation is permitted between wall heights and between factored axial loads. 

Depending on the reinforcement ratio and the eccentricity present, some economy 

is lost in using the table 4.5 values instead of the preceding calculation method.

TABLE 4.5 Simplified Moment Magnification Factors, δns

Minimum Wall 

Thickness (inches)

Maximum Wall Height 

(feet)

Factored Axial Load (plf)

2,000 4,000

5.5 8 1.07 1.15

10 1.12 1.26

7.5 8 1.03 1.06

10 1.04 1.09

9.5 8 1.00 1.03

10 1.00 1.04

Example 4.7 in section 4.9 presents the complete design of a reinforced 

concrete foundation wall. The magnified moment and corresponding total 

factored axial load are plotted on an interaction diagram as shown in figure 4.5 

(Refer to section 4.5.1.3 for a description of interaction diagrams and additional 

resources).

4.5.1.3 Interaction Diagrams

An interaction diagram is a graphic representation of the relationship 

between the axial load and bending capacity of a reinforced or plain concrete 

wall. The primary use of interaction diagrams is as a design aid for selecting 

predetermined concrete wall or column designs for varying loading conditions. 

Several publications provide interaction diagrams for use with concrete; however, 

these publications typically focus on column or wall design that is heavily 

reinforced, in accordance with design loads common in commercial construction. 

Residential concrete walls are either plain or slightly reinforced with one layer of 

reinforcement typically placed near the center of the wall. Plain and reinforced 

concrete interaction diagrams for residential applications and the methods for 

deriving them may be found in Structural Design of Insulating Concrete Form 

Walls in Residential Construction (PCA, 1998). StructurePoint, an affiliate of the 

Portland Cement Association (PCA) and the Cement Association of Canada, also 

offers a computer program that plots interaction diagrams based on user input; the 

program is titled spColumn (PCACOL).

An interaction diagram assists the designer in determining the wall’s 

structural adequacy under various loading conditions (in other words, 

combinations of axial and bending loads). Figure 4.5 illustrates interaction 

diagrams for plain and reinforced concrete. Both the design points located within 
Residential Structural Design Guide 4-33

the interaction curve for a given wall height and the reference axes represent a 

combination of axial load and bending moment that the wall can safely support. 

The most efficient design is close to the interaction diagram curve. For residential 

applications, the designer, realizing that the overall design process is not exact, 

may accept designs within plus or minus 5 percent of the interaction curve.

FIGURE 4.5 Typical Interaction Diagrams for Plain

and Reinforced Concrete Walls

Notes:

φPn = factored nominal load

φMn = factored nominal moment

Pn = nominal load

Mn = nominal moment

ft-kip = 1,000 ft-lb
Residential Structural Design Guide 4-34

4.5.1.4 Minimum Concrete Wall Reinforcement

Plain concrete foundation walls can provide serviceable structures when 

they are adequately designed (see section 4.5.1.1). When reinforcement is used to 

provide additional strength in thinner walls or to address more heavily loaded 

conditions, tests have shown that horizontal and vertical wall reinforcement 

spacing limited to a maximum of 48 inches on center results in performance that 

agrees reasonably well with design expectations (Roller, 1996). The designer 

should still ensure that the reinforcement area meets required minimum specified 

by the building code and that the reinforcement area is determined by acceptable 

methods. 

ACI 318•22.6.6.5 requires two No. 5 bars around all wall openings. The 

rebar, at a minimum, should be the same size required by the design of the 

reinforced wall, or a minimum No. 4 for plain concrete walls. In addition, a lintel 

(that is, concrete beam) is required at the top of wall openings; refer to section 

4.5.1.6 for more detail on lintels.

4.5.1.5 Concrete Wall Deflection

ACI 318 does not specifically limit wall deflection; therefore, deflection 

usually is not analyzed in residential foundation wall design. Regardless, a 

deflection limit of L/240 for unfactored soil loads is not unreasonable for belowgrade walls that are reinforced concrete. For plain concrete walls, such large 

deflections are not tolerable, and designing such walls for strength alone is 

considered to provide adequate rigidity and serviceability (refer to section 4.4). 

When using the moment magnifier method, the designer should apply the 

calculated moment magnification factor to the unfactored load moments used in 

conducting the deflection calculations. The calculation of wall deflection should 

also use effective section properties based on EcIg for plain concrete walls and 

EcIe for reinforced concrete walls; refer to ACI 318•9.5.2.3 to calculate the 

effective moment of inertia, Ie.

If unfactored load deflections prove unacceptable, the designer may 

increase the wall thickness or the amount of vertical wall reinforcement. For some 

residential loading conditions, however, satisfying reasonable deflection 

requirements should not be a limiting condition.

4.5.1.6 Concrete Wall Lintels

The loads over openings in concrete walls are supported by concrete, steel, 

precast concrete, cast stone, or reinforced masonry wall lintels. Wood headers 

also are used when not supporting concrete construction above and when 

continuity at the top of the wall (that is, a bond beam) is not critical, as in highhazard seismic or hurricane coastal zones, or is maintained sufficiently by a wood 

sill plate and other construction above.

This section focuses on the design of concrete lintels in accordance with 

chapters 10 and 11 of ACI 318. The concrete lintel often is assumed to act as a 
Residential Structural Design Guide 4-35

simple span, with each end pinned; however, the assumption implies no top 

reinforcement to transfer the moment developed at the end of the lintel. Under 

that condition, the lintel is assumed to be cracked at the ends, such that the end 

moment is zero, and the shear must be transferred from the lintel to the wall 

through the bottom reinforcement.

If the lintel is assumed to act as a fixed-end beam, the top and bottom 

reinforcement should be sufficiently embedded beyond each side of the opening 

to fully develop a moment-resisting end in the lintel. Although more complicated 

to design and construct, a fixed-end beam reduces the maximum bending moment 

(that is, wl2/12 instead of wl2/8) on the lintel and allows increased spans. A 

concrete lintel cast in a concrete wall acts somewhere between a true simple span 

beam and a fixed-end beam. Thus, a designer may design the bottom bar for a 

simple span condition and the top bar reinforcement for a fixed-end condition 

(conservative). Often, a No. 4 bar is placed at the top of each wall story to help tie 

the walls together (serving as a bond beam), which also can serve as the top 

reinforcement for concrete lintels. Figure 4.6 depicts the cross section and 

dimensions for analysis of concrete lintels. Example 4.5 demonstrates the design 

of a concrete lintel; refer to section 4.9.

For additional information on concrete lintels and their design procedure, 

refer to the Structural Design of Insulating Concrete Form Walls in Residential 

Construction (PCA, 1998) and to Testing and Design of Lintels Using Insulating 

Concrete Forms (HUD, 2000). The latter demonstrates, through testing, that shear 

reinforcement (that is, stirrups) of concrete lintels is not necessary for short spans 

(in other words, 3 feet or less) with lintel depths of 8 inches or more. This 

research also indicates that the minimum reinforcement requirements in ACI 318 

for beam design are conservative when a minimum No. 4 rebar is used as bottom 

reinforcement. Further, lintels with small span-to-depth ratios can be accurately 

designed as deep beams in accordance with ACI 318 when the minimum 

reinforcement ratios are met (Refer to ACI 318•11.4).
Residential Structural Design Guide 4-36

FIGURE 4.6 Design Variables Defined for Lintel Bending and Shear 

Flexural Capacity

The following equations are used to determine the flexural capacity of a 

reinforced concrete lintel, in conjunction with figure 4.6. An increase in the lintel 

depth or area of reinforcement is suggested if greater bending capacity is required. 

As a practical matter, though, lintel thickness is limited to the thickness of the 

wall in which a lintel is placed. In addition, lintel depth often is limited by the 

floor-to-floor height and the vertical placement of the opening in the wall. In 

many cases, therefore, increasing the amount or size of reinforcement is the most 

practical and economical solution.

ACI 318•10

Mu ≤ φ Mn

12

w M

2

u

 = for fixed-end beam model

8

w M

2

u

 = for simple span beam model





 



 φ = φ − 2

a Mn Asfy d

0.85f b

A f

a

c

s y

′ =

φ = 0.9
Residential Structural Design Guide 4-37

Shear Capacity

Concrete lintels are designed for shear resulting from wall, roof, and floor 

loads, in accordance with the equations below and figure 4.6.

ACI 318•11

Vu ≤ φVn

Vn = Vc + Vs

V 2 f b d c c w = ′

c w u c

v y

s 8 f b d when V V

s

A f d

V = ≤ ′ > φ

2

V when V

f

50b s A c u

y

w v,min

φ = >

{ or 24 in} 2

d s ≤ minimum of

{ or 12 in} when V 4 f b d 4

d s minimum of s c w ≤ > ′

φ = 0.85

Check Concrete Lintel Deflection

ACI 318 does not specifically limit lintel deflection; therefore, a 

reasonable deflection limit of L/240 for unfactored live loads is suggested. The 

selection of an appropriate deflection limit, however, is subject to designer 

discretion. In some applications, a lintel deflection limit of L/180 with live and 

dead loads is adequate. A primary consideration is whether the lintel is able to 

move independently of door and window frames. Calculation of lintel deflection 

should use unfactored loads and the effective section properties EcIe of the 

assumed concrete section (Refer to ACI 318•9.5.2.3 to calculate the effective 

moment of inertia Ie of the section).

4.5.2 Masonry Foundation Walls

Masonry foundation wall construction is common in residential 

construction. It is used in a variety of foundation types, including basements, 

crawl spaces, and slabs-on-grade. For prescriptive design of masonry foundation 

walls in typical residential applications, a designer or builder may use the IRC

(ICC, 2012) or the local residential building code.

ACI 530 develops methods for the design of masonry foundation walls by 

using allowable stress design; therefore, design loads may be determined 

according to load combinations presented in chapter 3 as follows:

• D + H

• D + H + 0.75 (Lr or S) + 0.75 L
Residential Structural Design Guide 4-38

In light-frame homes, the first load combination typically governs 

masonry walls for the same reasons stated in section 4.5.1 for concrete foundation 

walls. To simplify the calculations, the designer may conservatively assume that 

the wall story acts as a simple span with pinned ends, although such an 

assumption may tend to overpredict the stresses in the wall (for a discussion on 

calculating the loads on a structure, refer to chapter 3). Appendix A contains basic 

load diagrams and equations to assist the designer in calculating typical loading 

conditions and element-based structural actions encountered in residential design. 

Further, walls that are determined to have adequate strength to withstand shear 

and combined axial load and bending moment generally satisfy unspecified 

deflection requirements; therefore, foundation wall deflection is not discussed in 

this section. If desired, however, deflection may be considered as discussed in 

section 4.5.1.5 for concrete foundation walls.

To follow the design procedure, the designer must know the strength 

properties of various types and grades of masonry, mortar, and grout currently 

available on the market; section 4.2.2 discusses the material properties. With the 

loads and material properties known, the designer can then perform design checks 

for various stresses by following American Concrete Institute’s ACI 530 (ACI, 

2013). Residential construction rarely involves detailed masonry specifications 

but rather makes use of standard materials and methods familiar to local suppliers 

and trades.

An engineer’s inspection of a home is hardly ever required or requested 

under typical residential construction conditions. Inspection should be considered 

when masonry construction is specified in high-hazard seismic or hurricane-prone 

areas. ACI 530 makes no distinction between inspected and noninspected 

masonry walls and, therefore, does not require adjustments in allowable stresses 

based on level of inspection.

Residential designers should keep in mind that concrete masonry units 

(that is, block) are readily available in nominal 6-, 8-, 10-, and 12-inch 

thicknesses. It is generally more economical if the masonry unit compressive 

strength f'm ranges between 1,500 and 3,000 psi. The standard block used in 

residential and light commercial construction is usually rated at 1,900 psi. 

4.5.2.1 Unreinforced Masonry Design

ACI 530 addresses the design of unreinforced masonry to ensure that unit 

stresses and flexural stresses in the wall do not exceed certain maximum 

allowable stresses. ACI 530 provides for two methods of design: an empirical 

design approach and an ASD approach.

Walls may be designed in accordance with ACI 530•TMS 402 by using 

the empirical design method under the following conditions:

• The building is not located in Seismic Design Category D or E, as 

defined in NEHRP 2009 (FEMA, 2009) or ASCE 7-10 (that is, 

Seismic Zones 3 or 4 in most current and local building codes). (Refer 

to chapter 3.)
Residential Structural Design Guide 4-39

• Foundation walls do not exceed 8 feet in unsupported height.

• The distance between perpendicular vertical or horizontal supports for 

loadbearing masonry walls is a maximum of 18 times the wall 

thickness. This limit typically does not apply to residential basements 

as required in the IRC (ICC, 2012).

• Compressive stresses do not exceed the allowable stresses listed in 

ACI 530; compressive stresses are determined by dividing the design 

load by the gross cross-sectional area of the unit, per ACI 530.

• Backfill heights do not exceed those listed in table 4.5.

• Backfill material is nonexpansive and is tamped no more than 

necessary to prevent excessive settlement.

• Masonry is laid in running bond with Type M or S mortar.

• Lateral support is provided at the top of the foundation wall before 

backfilling.

Drainage is important when using the empirical table because lack of good 

drainage may substantially increase the lateral load on the foundation wall if the 

soil becomes saturated. As required in standard practice, the finish grade around 

the structure should be adequately sloped (minimum 1 inch of fall per foot of 

distance from the structure) to drain surface water away from the foundation 

walls. The backfill material should also be drained to remove ground water from 

poorly drained soils.

Out-of-plane bracing of the masonry foundation walls can be achieved by 

providing lateral support from the wood floor framing that is supported by and 

connected to the wall. The most common method of connection is a wood sill 

plate anchored to the top of the masonry wall with anchor bolts, and nailing of the 

floor framing to the sill plate (see chapter 7). Bracing by the floor system should 

be in place prior to the wall being backfilled.

When the limits of the empirical design method are exceeded, the ASD 

procedure for unreinforced masonry, as detailed herein, provides a more flexible 

approach by which walls are designed as compression and bending members, in 

accordance with ACI 530•2.2.

TABLE 4.6 Nominal Wall Thickness for 8-Foot-High Masonry

Foundation Walls1, 2

Nominal 

Wall 

Thickness

Maximum Unbalanced Backfill Height

Hollow Unit Masonry Solid Unit Masonry Fully Grouted Unit 

Masonry

6 inches 3 5 5

8 inches 5 5 7

10 inches 6 7 8

12 inches 7 7 8

Source: Modified from the ACI 530• 9.6 by using the IRC (ICC, 2012).

Notes:

1

Based on a backfill with an assumed equivalent fluid density of 30 pcf.
Residential Structural Design Guide 4-40

2

Backfill height is measured from the top of the basement slab to the finished exterior grade; wall height is measured from the top of the 

basement slab to the top of the wall.

The fundamental assumptions, derivation of formulas, and design 

procedures for ASD are similar to those developed for strength-based design for 

concrete except that the material properties of masonry are substituted for those of 

concrete. Allowable masonry stresses used in ASD are expressed in terms of a 

fraction of the specified compressive strength of the masonry at the age of 28 

days: f'm. A typical fraction of the specified compressive strength is 0.25 or 0.33, 

which equates to a conservative safety factor between 3 and 4 relative to the 

minimum specified masonry compressive strength. Table 4.7 provides design 

values for flexural tension stress. As in plain concrete, unreinforced masonry has 

very low tension capacity. The following design checks are used to determine if 

an unreinforced masonry wall is structurally adequate (refer to example 4.6 for 

the design of an unreinforced concrete masonry wall).

TABLE 4.7 Allowable Flexural Tension Stresses (Fa) for Allowable Stress

Design of Unreinforced Masonry

Type of Masonry Unit Construction

Mortar Type M or S

Portland Cement/Lime

(psi)

Masonry Cement and Air-Entrained 

Portland Cement/Lime

(psi)

Normal to Bed Joints

Solid 53 32

Hollow1

Ungrouted 33 20

Fully grouted 86 81

Parallel to Bed Joints in Running Bond

Solid 106 64

Hollow

Ungrouted/partially grouted 66 40

Fully grouted 106 64

Source: Table 2.2.3.2 TMS 402

Note:

1

For partially grouted masonry, allowable stresses may be determined on the basis of linear interpolation between fully grouted and 

ungrouted hollow units, based on the amount of grouting.

Shear Capacity

Shear stress is a result of the lateral loads on the structure associated with 

wind, earthquakes, or backfill forces. Lateral loads are both normal to the wall 

surface (that is, perpendicular or out of plane) and parallel to the wall surface (that 

is, parallel or in plane). Both perpendicular and parallel shear should be checked, 

as either could be a controlling factor in residential foundation walls.

If greater perpendicular shear capacity is required, it may be obtained by 

(1) increasing the wall thickness, (2) increasing the masonry unit compressive 

strength, or (3) adding vertical reinforcement in grouted cells. If greater parallel 
Residential Structural Design Guide 4-41

shear capacity is required, it may be obtained by (1) increasing the wall thickness, 

(2) reducing the size or numbers of wall openings, or (3) adding horizontal joint 

reinforcement. Horizontal truss-type joint reinforcement can substantially increase 

parallel shear capacity, provided that it is installed properly in the horizontal 

mortar bed joints. If not installed properly, it can create a place of weakness in the 

wall, particularly in out-of-plane bending of an unreinforced masonry wall.

The equations that follow are used to check perpendicular and parallel 

shear in masonry walls. The variable Nv is the axial design load acting on the wall 

at the point of maximum shear. The equations are based on An, which is the net 

cross-sectional area of the masonry. For parallel shear, the equations do not 

address overturning and bending action that occurs in a direction parallel to the 

wall, particularly for short segments of walls under significant parallel shear load. 

ACI 530•2.2.5

v Fv f ≤

n

v 2A

3V f =

Fv = minimum of 















+ for running bond

A

N

37 psi 0.45

120psi

1.5 f' for axial and shear members

n

v

m

Axial Compression Capacity

The following equations from ACI 530•2.3 are used to design masonry 

walls and columns for compressive loads only. They are based on the net crosssectional area of the masonry, including grouted and mortared areas.

ACI 530•2.3

Columns

P ≤ Pa

( ) 99

r

where h

140r

h P 0.25f A 0.65A F 1

2

a m n st s ≤















 



 



 = ′ + −

( ) 99

r

where h

h

70r P 0.25f A 0.65A F

2

a m n st s  > 

 



 = ′ +

Pa,maximum = FaAn

An

I

r =

Walls

a Fa f ≤

( ) 99

r

where h

140r

h F 0.25f 1

2

a m ≤





















 



 = ′ −
Residential Structural Design Guide 4-42

ACI 530•2.3

A f P a =

( ) 99

r

where h

h

70r F 0.25f

2

a m  > 

 



 = ′

12

t

A

I

r

n

= ≅

3

2

m

2

e )

r

e (1 0.577

h

TT E I

P = −

P < 1/4 Pe

Em = 900 F'm

Combined Axial Compression and Flexural Capacity

The following equations from ACI 530 determine the relationship of the 

combined effects of axial load and bending moment on a masonry wall.

ACI 530•2.3

1

F

f

F

f

b

b

a

a + ≤

n

a A

P f =

25Pe P ≤ 0.

( ) 99

r

forh

140r

h F 0.25f 1

2

a m ≤





















 



 = ′ −

( ) 99 r forh

h

70r F 0.25f

2

a m  > 

 



 = ′

An

I

r =

S

M f b =

Fb 0.33f m = ′

3

2

m

2

e r

e 1 0.577

h

E I

P 



 



 − π =

E m 900f m = ′

ft < Ft

Ft = ACI 530 Table 2.2.3.2

S

M

A

P f

n

t + − =
Residential Structural Design Guide 4-43

Tension Capacity

ACI 530 provides allowable values for flexural tension transverse to the 

plane of a masonry wall. Standard principles of engineering mechanics determine 

the tension stress resulting from the bending moment caused by lateral (that is, 

soil) loads and offset by axial (that is, dead) loads.

ACI 530•2.3

ft < Ft

Ft = ACI 530 Table 2.2.3.2

S

M

A

P f

n

t = +

Even though an unreinforced masonry wall may calculate as adequate, the 

designer may consider adding a nominal amount of reinforcement to control 

cracking (Refer to section 4.5.2.3 for a discussion on nominal reinforcement).

Walls determined inadequate to withstand combined axial load and 

bending moment may gain greater capacity through (1) increased wall thickness, 

(2) increased masonry compressive strength, or (3) the addition of steel 

reinforcement. Usually the most effective and economical solution for providing 

greater wall capacity in residential construction is to increase wall thickness, 

although reinforcement also is common. Section 4.5.2.2 discusses the design 

procedure for a reinforced masonry wall.

4.5.2.2 Reinforced Masonry Design

When unreinforced concrete masonry wall construction does not satisfy all 

design criteria (load, wall thickness limits, and so on), reinforced walls may be 

designed by following the ASD procedure or the strength-based design procedure 

of ACI 530. The ASD procedure outlined herein describes an approach by which 

walls are designed in accordance with ACI 530•2.3. Although not discussed in 

detail herein, walls may also be designed by following the strength-based design 

method specified in ACI 530.

For walls designed in accordance with ACI 530•2.3 using the ASD 

method, the fundamental assumptions, derivation of formulas, and design 

procedures are similar to those for design using concrete except that the material 

properties of masonry are substituted for those of concrete. Allowable masonry 

stresses used in ASD are expressed in terms of a fraction of the specified 

compressive strength of the masonry at the age of 28 days, f'm. A typical fraction 

of the specified compressive strength is 0.25, which equates to a conservative 

safety factor of 4. The following design checks determine whether a reinforced 

masonry wall is structurally adequate (refer to example 4.7 for the design of a 

reinforced concrete masonry wall).
Residential Structural Design Guide 4-44

Shear Capacity

Shear stress is a result of lateral loads on the structure associated with 

wind, earthquakes, or backfill forces. Lateral loads are both normal to the wall 

surface (that is, perpendicular or out of plane) and parallel to the wall surface (that 

is, parallel or in plane). Both perpendicular and parallel shear should be checked; 

perpendicular shear may be a controlling factor in the design of masonry walls, 

and parallel shear could be a controlling factor if the foundation is partially or 

fully above grade (such as a walkout basement) with a large number of openings. 

The equations that follow check perpendicular and parallel shear in 

conjunction with figure 4.7. Some building codes include a “j” coefficient in these 

equations. The “j” coefficient defines the distance between the center of the 

compression area and the center of the tensile steel area; however, it often is 

dismissed or approximated as 0.9. If greater parallel shear capacity is required, it 

may be obtained in a manner similar to that recommended in the previous section 

for unreinforced masonry design. For parallel shear, the equations do not address 

overturning and bending action that occurs in a direction parallel to the wall, 

particularly for short segments of walls under significant parallel shear load. 

ACI 530•7.5

v Fv f ≤

fv = V

Anv

F 1.0 f 50 psi v = m

′ ≤ for flexural members

psi Vd

M f 80 45

Vd

M 4

3

1 Fv m 



 



  ′ ≤ − 

 



 = −

 for shear walls where 1

Vd

M

<

F 1.0 f 35 psi v = m

′ ≤

 for shear walls where 1

Vd

M

≥

If the shear stress exceeds the above allowables for masonry only, the 

designer must design shear reinforcing with the shear stress equation changes, in 

accordance with ACI 530•2.3.5. In residential construction, increasing the wall 

thickness or grouting additional cores is generally more economical than using 

shear reinforcement. If shear reinforcement is desired, refer to ACI 530. ACI 530 

limits vertical reinforcement to a maximum spacing (s) of 48 inches. Flexural or 

axial stresses must be accounted for to ensure that a wall is structurally sound. 

Axial loads increase compressive stresses and reduce tension stresses and may be 

great enough to keep the masonry in an uncracked state under a simultaneous 

bending load. 
Residential Structural Design Guide 4-45

Axial Compression Capacity

The following equations from ACI 530•2.3 are used to determine whether 

a masonry wall can withstand conditions when compressive loads act only on 

walls and columns (that is, interior load-bearing wall or floor beam support pier). 

As with concrete, compressive capacity usually is not an issue in supporting a 

typical light-frame home. An exception may occur with the bearing points of 

long-spanning beams. In such a case, the designer should check bearing capacity 

by using ACI 530•2.1.7.
Residential Structural Design Guide 4-46

FIGURE 4.7 Shear Calculations in Reinforced

Concrete Masonry Walls: Variables Defined

Notes:

d = distance to neutral axis

L = length

Sv = perpendicular shear area

t = thickness

v = perpendicular shear
Residential Structural Design Guide 4-47

ACI 530•2.3

Columns

P ≤ Pa

( ) 99

r

where h

140r

h P 0.25f A 0.65A F 1

2

a m n st s ≤





















 



 = ′ + −

( ) 99

r

where h

h

70r P 0.25f A 0.65A F

2

a m n st s  > 

 



 = ′ +

Pa,maximum = FaAn

Ae

I

r =

Walls

a Fa f ≤

( ) 99

r

where h

140r

h F 0.25f 1

2

a m ≤





















 



 = ′ −

( ) 99

r

where h

h

70r F 0.25f

2

a m  > 

 



 = ′

Ae

I

r =

Calculation using the preceding equations is based on Ae, which is the 

effective cross-sectional area of the masonry, including grouted and mortared 

areas substituted for An.

Combined Axial Compression and Flexural Capacity

In accordance with ACI 530•2.3.2, the design tensile forces in the 

reinforcement due to flexure shall not exceed 20,000 psi for grade 40 or 50 steel, 

24,000 psi for grade 60 steel, or 30,000 psi for wire joint reinforcement. As stated, 

most reinforcing steel in the U.S. market today is grade 60. The following equations 

pertain to walls that are subject to combined axial and flexure stresses.

ACI 530•7.3

Fb 0.33f m = ′

b

a

a b F

F

f

1

S

M f 













 = ≤ −

Columns

1

F

f

P

P

b

b

a

+ ≤
Residential Structural Design Guide 4-48

ACI 530•7.3

( ) 99

r

where h

140r

h P 0.25f A 0.65A F 1

2

a m n st s ≤















 



 



 = ′ + −

( ) 99

r

where h

h

70r P 0.25f A 0.65A F

2

a m n st s  > 

 



 = ′ +

Walls

1

F

f

F

f

b

b

a

a + ≤

m

e

a 0.33f

A

P f = ≤ ′ due to flexure only or flexure in combination with axial load

( ) 99

r

forh

140r

h F 0.25f 1

2

a m ≤















 



 



 = ′ −

( ) 99 r forh

h

70r F 0.25f

2

a m  > 

 



 = ′

Walls determined inadequate to withstand combined axial load and bending

moment may gain greater capacity through (1) increased wall thickness, (2) 

increased masonry compressive strength, or (3) added steel reinforcement.

4.5.2.3 Minimum Masonry Wall Reinforcement

For reinforced concrete masonry shear walls, ACI 530 stipulates minimum 

reinforcement limits as shown herein. The designer should rely on experience in 

local practice and local building code provisions for prescriptive masonry 

foundation or above-grade wall design in residential applications.

ACI 530•2.3.5

F d

M A

s

s,required =

4.5.2.4 Masonry Wall Lintels

Openings in masonry walls are constructed by using steel, precast 

concrete, or reinforced masonry lintels. Wood headers also are used when they do 

not support masonry construction above and when continuity at the top of the wall 

(a bond beam) is not required or is adequately provided within the system of 

wood-framed construction above. Steel angles are the simplest shapes and are 

suitable for openings of moderate width typically found in residential foundation 

walls. The angle should have a horizontal leg of the same width as the thickness 

of the concrete masonry that it supports. Openings may require vertical 

reinforcing bars with a hooked end that is placed on each side of the opening to 

restrain the lintel against uplift forces in high-hazard wind or earthquake regions. 

Building codes typically require steel lintels exposed to the exterior to be a 
Residential Structural Design Guide 4-49

minimum 1/4-inch thick. Figure 4.8 illustrates some lintels commonly used in 

residential masonry construction.

FIGURE 4.8 Concrete Masonry Wall Lintel Types
Residential Structural Design Guide 4-50

Many prescriptive design tables are available for lintel design. For more 

information on lintels, arches, and their design, refer to the National Concrete 

Masonry Association’s (NCMA’s) TEK Notes. Information on lintels and arches 

also can be found in Masonry Design and Detailing (Beall, 2012).

4.5.3 Preservative-Treated Wood Foundation Walls

Preservative-treated wood foundations, commonly known as permanent 

wood foundations (PWF), have been used in more than 300,000 homes and other 

structures throughout the United States. When properly installed, they provide 

foundation walls at an affordable cost. In some cases, the manufacturer may offer 

a 50-year material warranty, which exceeds the warranty offered for other 

common foundation materials.

A PWF is a load-bearing, preservative-treated, wood-framed foundation 

wall sheathed with preservative-treated plywood; it bears on a gravel spread 

footing. PWF lumber and plywood used in foundations are pressure treated with 

chromated copper arsenate (CCA) or other approved preservatives (AWPA, 

2013). The walls are supported laterally at the top by the floor system and at the 

bottom by a cast-in-place concrete slab, a pressure-treated lumber floor system, or 

backfill on the inside of the wall. Proper connection details are essential, along 

with provisions for drainage and moisture protection. All fasteners and hardware 

used in a PWF should be stainless steel or hot-dipped galvanized steel. Figure 4.9 

illustrates a PWF.

PWFs may be designed in accordance with the basic provisions in the IRC

(ICC, 2012). Those provisions, in turn, are based on the American Forest and 

Paper Association’s Permanent Wood Foundation Design Specification (AF&PA, 

2007). The PWF guide offers design flexibility and thorough technical guidance. 

Table 4.7 summarizes some basic rules of thumb for design, and the steps for 

using the prescriptive tables follow.
Residential Structural Design Guide 4-51

FIGURE 4.9 Preservative-Treated Wood Foundation Walls
Residential Structural Design Guide 4-52

TABLE 4.8 Preservative-Treated Wood Foundation Framing1

Maximum Unbalanced 

Backfill Height (feet) Nominal Stud Size Stud Center-to-Center Spacing(inches)

5 2x6 16

6 2x6 12

8 2x8 12

• Connect each stud to top plate with framing anchors when the backfill height is 6 feet or greater.

• Provide full-depth blocking in the outer joist space along the foundation wall when floor joists are oriented 

parallel to the foundation wall.

• The bottom edge of the foundation studs should bear against a minimum of 2 inches of the perimeter 

screed board or the basement floor to resist shear forces from the backfill.

1

Connection of studs to plates and plates to floor framing is critical to the performance of permanent wood foundations. The building code 

and the Permanent Wood Foundation Design Specification (AF&PA, 2007) should be carefully consulted with respect to connections.

• Granular (that is, gravel or crushed rock) footings are sized in 

accordance with section 4.4.1. Permanent wood foundations may also 

be placed on poured concrete footings.

• Footing plate size is determined by the vertical load from the structure 

on the foundation wall and the size of the permanent wood foundation 

studs.

• The size and spacing of the wall framing is selected from tables for 

buildings up to 36 feet wide that support one or two stories above 

grade.

• APA-rated plywood is selected from tables based on unbalanced 

backfill height and stud spacing. The plywood must be treated with 

preservatives and rated for below-ground application.

• Drainage systems are selected in accordance with foundation type (for 

example, basement or crawl space) and soil type. Foundation wall 

moisture proofing (that is, polyethylene sheeting) also is required.

For more information on preservative-treated wood foundations and their 

specific design and construction, consult the Permanent Wood Foundation 

Specification (AF&PA, 2007).

4.5.4 Insulating Concrete Form Foundation Walls

Insulating concrete forms (ICFs) have been used in the United States since 

the 1970s. They provide durable and thermally efficient foundation and abovegrade walls at reasonable cost. ICFs are constructed of rigid foam plastic, 

composites of cement and plastic foam insulation or wood chips, or other suitable 

insulating materials that have the ability to act as forms for cast-in-place concrete 

walls. The forms are easily placed by hand and remain in place after the concrete 

is cured to provide added insulation.
Residential Structural Design Guide 4-53

ICF systems typically are categorized with respect to the form of the ICF 

unit. There are three types of ICF forms: hollow blocks, planks, and panels. The 

shape of the concrete wall is best visualized with the form stripped away, 

exposing the concrete to view. Following are the ICF categories based on the 

resulting nature of the concrete wall.

• Flat. Solid concrete wall of uniform thickness.

• Post-and-beam. Concrete frame constructed of vertical and horizontal 

concrete members with voids between the members created by the 

form. The spacing of the vertical members may be as great as 8 feet.

• Screen-grid. Concrete wall composed of closely spaced vertical and 

horizontal concrete members with voids between the members created 

by the form. The wall resembles a thick screen made of concrete.

• Waffle-grid. Concrete wall composed of closely spaced vertical and 

horizontal concrete members with thin concrete webs filling the space 

between the members. The wall resembles a large waffle made of 

concrete.

Foundations may be designed in accordance with the values provided in 

the most recent national building codes’ prescriptive tables (ICC, 2012). 

Manufacturers also usually provide design and construction information. ICF 

walls are designed by following a procedure similar to that in section 4.5.1; 

however, special consideration must be given to the dimensions and shape of an 

ICF wall that is not a flat concrete wall (refer to figure 4.10 for a typical ICF 

foundation wall detail).
Residential Structural Design Guide 4-54

FIGURE 4.10 Insulating Concrete Form Foundation Walls

For more design information, consult the Prescriptive Design of Exterior 

Concrete Walls for One- and Two-Family Dwellings (PCA-100, 2007) or the 

Prescriptive Method for Insulating Concrete Forms in Residential Construction

(HUD, 2002). 

4.6 Slabs on Grade

The primary objectives of slab-on-grade design are—

• To provide a floor surface with adequate capacity to support all 

applied loads.

• To provide thickened footings for attachment of the above-grade 

structure and for transfer of the load to the earth where required.

• To provide a moisture barrier between the earth and the interior of the 

building.

Many concrete slabs for homes, driveways, garages, and sidewalks are 

built according to standard thickness recommendations and do not require a 
Residential Structural Design Guide 4-55

specific design unless poor soil conditions, such as expansive clay soils, exist on 

the site. 

For typical loading and soil conditions, floor slabs, driveways, garage 

floors, and residential sidewalks are built at a nominal 4 inches thick per ACI 

302•2.1. Where interior columns and load-bearing walls bear on the slab, the slab 

typically is thickened and may be nominally reinforced (refer to section 4.4 for 

footing design procedures). Monolithic slabs may also have thickened edges that 

provide a footing for structural loads from exterior load-bearing walls. The 

thickened edges may or may not be reinforced in accordance with the loads and 

the soil-bearing capacity.

Slab-on-grade foundations often are placed on 2 to 3 inches of washed 

gravel or sand and a 6-mil (0.006 inch) polyethylene vapor barrier. This 

recommended practice prevents moisture in the soil from wicking through the 

slab. The sand or gravel layer acts primarily as a capillary break to soil moisture 

transport through the soil. If tied into the foundation drain system, the gravel layer 

also can help provide drainage.

A slab-on-grade greater than 10 feet in any dimension will likely 

experience cracking from temperature and shrinkage effects that create internal 

tensile stresses in the concrete. To prevent the cracks from becoming noticeable, 

the designer usually specifies reinforcement, such as welded wire fabric (WWF) 

or a fiber-reinforced concrete mix. The location of cracking may be controlled by 

placing construction joints in the slab at regular intervals or at strategic locations 

hidden under partitions or under certain floor finishes (that is, carpet).

In poor soils in which reinforcement is required to increase the slab’s 

flexural capacity, the designer should follow conventional reinforced concrete 

design methods. The Portland Cement Association, Wire Reinforcement Institute 

(WRI), and U.S. Army Corps of Engineers (USACE) each espouse a different

method for the design of plain or reinforced concrete slabs-on-grade.

Presented in chart or tabular format, the PCA method selects a slab 

thickness in accordance with the applied loads and is based on the concept of one 

equivalent wheel loading at the center of the slab. PCA design typically does not 

require structural reinforcement; however, a nominal amount of reinforcement is 

suggested for minimizing cracks, shrinkage, and temperature effects.

The WRI method selects a slab thickness in accordance with a discreteelement computer model for the slab. The approach graphically accounts for the 

relative stiffness between grade support and the concrete slab to determine 

moments in the slab and presents the information in the form of design 

nomographs.

Presented in charts and tabular format, the USACE method is based on 

Westergaard’s (1926) formulae for edge stresses in a concrete slab. This method 

assumes that the unloaded portions of the slab help support the slab portions 

under direct loading.

For further information on the design procedures for each design method 

mentioned and for unique loading conditions, refer to ACI 360, Guide to Design

of Slabs on Ground (ACI, 2010), or Design and Construction of Post-Tensioned 

Slabs on Ground (PTI, 2008) for expansive soil conditions.
Residential Structural Design Guide 4-56

4.7 Pile Foundations

Piles support buildings under a variety of special conditions that make 

conventional foundation practices impractical or inadvisable. Such conditions 

include—

• Weak soils or nonengineered fills that require the use of piles to 

transfer foundation loads by skin friction or point bearing.

• Inland floodplains and coastal flood hazard zones where buildings 

must be elevated.

• Steep or unstable slopes.

• Expansive soils on which buildings must be isolated from soil 

expansion in the “active” surface layer and anchored to stable soil 

below.

Piles are available in a variety of materials and different mechanisms of 

support. Preservative-treated timber piles typically are driven into place by a 

crane with a mechanical or drop hammer (most common in weak soils and coastal 

construction). Concrete piles or piers typically are cast in place in drilled holes, 

sometimes with “belled” bases (most common in expansive soils). Steel H-piles 

or large-diameter pipes are typically driven or vibrated into place with specialized 

heavy equipment (uncommon in residential construction). Helical piles have 

screw flights on the end that are “screwed” into the ground until they reach 

refusal. They most often terminate in a concrete grade beam to tie the tops of the 

piles together, thereby reducing lateral movement of the foundation system.

Timber piles most commonly are used in light-frame residential 

construction. The minimum pile capacity is based on the required foundation 

loading. Pile capacity is, however, difficult to predict; therefore, designers are 

able to make only rough estimates of required pile lengths and sizes before 

installation, particularly when the designer relies only on skin friction to develop 

capacity in deep, soft soils. For this reason, being familiar with local successful 

practice is a factor in any pile foundation design. A pile foundation sometimes can 

be specified by drawing on experience, with minimal design effort, in locations 

not subject to flooding or other extreme loadings from high winds or earthquakes. 

In other cases, some amount of subsurface exploration (that is, by using a 

standard penetrometer test) is advisable to assist in foundation design or, 

alternatively, to indicate when one or more test piles may be required.

Pile depth rarely has to be greater than 8 or 10 feet except in extremely 

soft soils, on steeply sloped sites with unstable soils, or in coastal hazard areas 

(that is, beachfront property) where significant scour is possible from storm surge 

velocity. Under these conditions, depths can easily exceed 15 feet and often reach 

25 feet. In coastal high-hazard areas known as “V zones” on flood insurance 

rating maps (FIRMs), the building must be elevated above the 100-year flood 

elevation, which is known as the base flood elevation (BFE) and includes an 
Residential Structural Design Guide 4-57

allowance for wave height. Figure 4.11 shows how treated timber piles typically 

are used to elevate a structure.

FIGURE 4.11 Basic Coastal Foundation Construction

For additional guidance, the designer should refer to the Coastal Construction 

Manual, FEMA P-55 (FEMA, 2011a) and Home Builder’s Guide to Coastal 

Construction, FEMA P-499 (FEMA, 2011b), both of which are updated 

frequently by the Federal Emergency Management Agency (FEMA). Another 

helpful resource is Pile Driving by Pile Buck (Pile Buck, 2011). Of course, 

designers should be prepared to make reasonable design modifications and 

judgments based on personal experience with and knowledge of pile construction 

and local conditions. The designer should also carefully consider National Flood 

Insurance Program (NFIP) requirements because they may affect the availability 

and cost of insurance. From a life-safety perspective, people often evacuate pilesupported buildings during a major hurricane, but flood damage can be substantial 

if the building is not properly elevated and detailed. In these conditions, the 

designer must consider several factors, including flood loads, wind loads, scour, 

breakaway wall and slab construction, corrosion, and other factors. 

The habitable portion of buildings in coastal “A zones” (nonvelocity flow) 

and inland floodplains must be elevated above the BFE, particularly if owners 
Residential Structural Design Guide 4-58

will be seeking flood insurance. Piles or other forms of an open foundation are the 

recommended method for constructing a foundation in coastal “A zones.”

The designer must specify a required minimum penetration length and the 

required axial capacity so the installer can equate driving resistance to sufficient 

bearing capacity. The designer should use pile capacity formulas such as those 

provided by the Navy guide titled Foundations and Earth Structures, Design 

Manual 7.02 (NAVFAC, 1986). The pile size may be specified as a minimum tip 

diameter, a minimum butt diameter, or both. The minimum pile butt diameter 

should be no less than 8 inches; 10- to 12-inch diameters are common. The larger 

pile diameters may be necessary for unbraced conditions with long, unsupported 

heights.

In hard material or densely compacted sand or hard clay, a typical pile 

meets “refusal” when the blows per foot become excessive. In such a case, the 

builder may need to jet or predrill the pile to a specific depth to meet the 

minimum embedment and then finish with several hammer blows to ensure that 

the required capacity is met and the pile properly seated in firm soil. When using 

either jetting or drilling as an installation method, the designer must consider 

reducing the capacity of the pile.

Jetting is the process of using a water pump, hose, and long pipe to “jet” 

the tip of the pile into hard-driving ground, such as firm sand. Jetting may also be 

used to adjust the pile vertically to maintain a reasonable tolerance with the 

building layout dimension.

Connecting or anchoring the building properly to pile foundations is 

important when severe uplift or lateral load conditions are expected. For standard 

pile and concrete grade beam construction, the pile is usually extended into the 

concrete “cap” a few inches or more. The connection requirements of the 

National Design Specification for Wood Construction (NDS; AWC, 2012) should 

be carefully followed for these heavy-duty connections. Such connections are not 

specifically addressed in chapter 7, although much of the information in that 

chapter is applicable to the topic.

4.8 Frost Protection

The objective of frost protection in foundation design is to prevent damage 

to the structure from frost action (that is, heaving and thaw weakening) in frostsusceptible soils.

4.8.1 Conventional Methods

In northern U.S. climates, builders and designers mitigate the effects of 

frost heave by constructing homes with perimeter footings that extend below a 

locally prescribed frost depth. Other construction methods include—

• Piles or caissons extending below the seasonal frost line.
Residential Structural Design Guide 4-59

• Mat or reinforced structural slab foundations that resist differential 

heave.

• Non-frost-susceptible fills and drainage. 

• Adjustable foundation supports.

The local building department typically sets required frost depths. Often, 

the depths set for residential foundations are highly conservative compared with 

frost depths relevant to other applications. The local design frost depth can vary 

significantly from that required by actual climate, soil, and application conditions. 

One exception occurs in Alaska, where it is common to specify different frost 

depths for “warm,” “cold,” and “interior” foundations. For homes in the 

Anchorage, Alaska, area, the perimeter foundation generally is classified as warm, 

with a required frost depth of 4 to 5 feet. Interior footings may be required to be 8 

inches deep. On the other hand, frost depth requirements for cold foundations, 

including outside columns, may be as much as 10 feet. In the contiguous 48 

states, depths for footings range from a minimum 12 inches in the South to as 

much as 6 feet or more in some northern localities.

Based on the air-freezing index, table 4.8 presents minimum “safe” frost 

depths for residential foundations. Figure 4.12 depicts the air-freezing index, a 

climate index closely associated with ground freezing depth. The most frostsusceptible soils are silty soils, or mixtures that contain a large fraction of siltsized particles. Generally, soils or fill materials with less than 6 percent fines (as 

measured by a #200 sieve) are considered non-frost-susceptible. Proper surface 

water and foundation drainage also are important factors where frost heave is a 

concern. The designer should recognize that many soils may not be frost 

susceptible in their natural state (such as sand, gravel, or other well-drained soils 

that are typically low in moisture content). For those soils that are frost 

susceptible, however, the consequences can be significant and costly if not 

properly considered in the foundation design.

TABLE 4.9 Minimum Frost Depths for Residential Footings1, 2

Air-Freezing Index (°F-Days) Footing Depth (inches)

250 or less 12

500 18

1,000 24

2,000 36

3,000 48

4,000 60

1

Interpolation is permissible.

2

The values do not apply to mountainous terrain or to Alaska.
Residential Structural Design Guide 4-60

4.8.2 Frost-Protected Shallow Foundations

A frost-protected shallow foundation (FPSF) is a practical alternative to 

deeper foundations in cold regions characterized by seasonal ground freezing and 

the potential for frost heave. Figure 4.13 illustrates several FPSF applications. 

FPSFs are best suited to slab-on-grade homes on relatively flat sites. The FPSF 

method may be used effectively with walkout basements, however, by insulating 

the foundation on the downhill side of the house, thus eliminating the need for a 

stepped footing

An FPSF is constructed by using strategically placed vertical and 

horizontal insulation to insulate the footings around the building, thereby allowing 

foundation depths as shallow as 12 inches in very cold climates. FPSF technology 

recognizes earth as a heat source that repels frost. Heat input to the ground from 

buildings therefore contributes to the thermal environment around the foundation.

The thickness of the insulation and the horizontal distance that the 

insulation must extend away from the building depends primarily on the climate. 

In less severe cold climates, horizontal insulation is not necessary. Other factors 

such as soil thermal conductivity, soil moisture content, and the internal 

temperature of a building are also important determinants of insulation use. 

Current design and construction guidelines are based on reasonable worst-case 

conditions.

After more than 40 years of use in the Scandinavian countries, FPSFs are 

now recognized in the prescriptive requirements of the IRC (ICC, 2012);

however, the code places limits on the use of foam plastic below grade in areas of 

noticeably high termite infestation probability. In those areas, termite barriers or 

other modifications must be incorporated into the design to block “hidden” 

pathways leading from the soil into the structure between the foam insulation and 

the foundation wall. The exception to the code limit occurs when termite-resistant 

materials (for example, concrete, steel, or preservative-treated wood) are specified 

for a home’s structural members.
Residential Structural Design Guide 4-61

FIGURE 4.12 Air-Freezing Index Map (100-Year Return Period)

Source: Reprinted with permission from the International Code Council, Washington, D.C. Copyright ICC.

Note: The air-freezing index is defined as the number of cumulative degree days below 32° F and is a measure of the magnitude and 

duration of below freezing air temperatures.
Residential Structural Design Guide 4-62

The complete design procedure for FPSFs is detailed in Frost Protected 

Shallow Foundations in Residential Construction, Second Edition (NAHB, 1996). 

The first edition of this guide is available from the U.S. Department of Housing 

and Urban Development. Either version provides useful construction details and 

guidelines for determining the amount (thickness) of insulation required for a 

given climate or application. Acceptable insulation materials include expanded 

and extruded polystyrenes, although adjusted insulation values are provided for 

below-ground use.

The American Society of Civil Engineers also has a standard for FPSF 

design and construction based on the resources mentioned. This standard is titled 

Design Guide for Frost-Protected Shallow Foundations, ASCE 32-01 (ASCE,

2001).
Residential Structural Design Guide 4-63

FIGURE 4.13 Frost-Protected Shallow Foundation Applications
Residential Structural Design Guide 4-64

4.9 Design Examples

EXAMPLE 4.1 Plain Concrete Footing Design

Given Exterior continuous wall footing supporting an 8-inch-wide concrete foundation wall 

carrying a 12-foot floor tributary width; the wall supports two floor levels, each with the 

same tributary width.

Design Loads

Live load 0.75 [(12 ft)(40 psf) +(12 ft)(30 psf)] = 630 plf (Table 3.1)

Dead load (12 ft)(10 psf)(2 floors) = 240 plf (Table 3.2)

Wall dead load (8 ft)(0.66 ft)(150 pcf) = 800 plf (Table 3.3)

Footing dead load allowance = 200 plf

Presumptive soil-bearing capacity = 1,500 psf (default)

f'c = 2,000 psi

Find The minimum size of the concrete footing required to support the loads

Solution

1. Determine the required soil-bearing area.

( )( ) ft plf ft 1.25

1,500psf

630 240 800 200 1

Soil bearing

Design load Footing width = + + + = =

The required footing width is equal to 

b = 1.25 ft = 15 in ≅ 16 in (standard width of excavation equipment)

2. Preliminary design (rule-of-thumb method)

Footing projection = 1/2 (16 in - 8 in) = 4 in

Required plain concrete footing thickness ≅ 4 in (no less than the projection)

∴ Use minimum 6-inch-thick footing.

Footing weight = (1.33 ft)(0.5 ft)(150 pcf) = 100 lb < 200 lb allowance OK

3.

Consider design options.

• Use 6-inch x 16-inch plain wall concrete footing.

• Design plain concrete footing to check rule of thumb for illustrative purposes 

only.
Residential Structural Design Guide 4-65

Design a plain concrete footing.

(a) Determine soil pressure based on factored loads.

( )( ) ( )( ) 1,877 psf (1.33ft)(1 ft)

1.2 240 plf 800plf 200plf 1.6 630plf

A

P

q

footing

u s = + + + = =

(b) Determine thickness of footing based on moment at the face of the wall.

( ) s 2

u b T

8

q M = − 

( )( ) (1.33ft 0.66 ft) 105 ft lb / lf

8

1,877 psf 1 ft 2 = − = −

6

b t

M 5 f' S 5 2,000 psi

2

φ n = c =

φ Mn ≥ Mu

( )( ) ( )( )( ) ( ) 













 − ≥

6

12 in t 105 ft lb / lf 12 in / ft 0.65 5 2,000 psi

2

t = 2.1 in

(c) Determine footing thickness based on one-way (beam) shear. 

f t

3

4 Vc c φ = φ ′ 

2,000psi (12in)(t) 3

4 0.65 



 



 =

V (q )(0.5 (b T) t) u = s − −

= (1,849 psf )(1ft)(0.5 (1.33ft − 0.66 ft)− t)

φVc ≥ Vu

2,000 psi (12in)(t) (1,877 psf )(1ft)(0.5(1.33ft 0.66 ft) t) 3

4 0.65  = − − 

 





t = 0.27 ft = 3.2 in

Therefore, shear in the footing governs the footing thickness.

Conclusion The calculations yield a footing thickness of 3.2 inches. In accordance with ACI 

318•22.4.8, two additional inches must be added, resulting in a footing thickness of 5.2 

inches. In accordance with ACI 318•22.7.4, however, plain concrete footings may not 

have a thickness less than 8 inches. In this case, a more economical and code-compliant 

footing design (6 inches thick) can be achieved by following the IRC prescriptive 

provisions for footings rather than following ACI provisions.
Residential Structural Design Guide 4-66

In high-hazard seismic areas, a nominal footing reinforcement should be considered (for 

example, one No. 4 bar longitudinally); however, longitudinal reinforcement at the top 

and bottom of the foundation wall provides greater strength against differential soil 

movement in a severe seismic event, particularly on sites with soft soils.
Residential Structural Design Guide 4-67

EXAMPLE 4.2 Reinforced Footing Design

Given Interior footing supporting a steel pipe column (3.5 in x 3.5 in bearing) carrying a 12-ft x 

12-ft floor tributary area

Service Loads

Live load (12 ft)(12 ft)(40 psf) = 5,760 lb

Dead load (12 ft)(12 ft)(10 psf) = 1,440 lb

Footing and column dead load = 300 lb (allowance)

Presumptive soil bearing capacity = 1,500 psf (default)

f'c = 2,500 psi, fy = 60,000 psi

Find The minimum size of the concrete footing required to support the loads

Solution

1. Determine the required soil-bearing area.

Area required = Service load

Presumptive soil bearing = (5,760 lb + 1,440 lb + 300 lb)1,500 psf = 5 ft 2

Assume a square footing

2. Preliminary design (rule-of-thumb method)

Footing projection = 1/2 (26 in - 3.5 in) = 11.25 in

∴ Required plain concrete footing thickness ≅ 12 in

Footing weight = (5 ft2

)(1 ft)(150 pcf) = 750 lb > 300 lb allowance

∴ Recalculation yields a 28-in x 28-in footing.

3. Consider design options.

• Use 12-in x 28-in x 28-in plain concrete footing (5 ft3 of concrete per footing, less 

expensive).

• Reduce floor column spacing (more but smaller footings, perhaps smaller floor 

beams, more labor).

• Test soil bearing to see if higher bearing value is feasible (uncertain benefits, but 

potentially large, perhaps one-half reduction in plain concrete footing size).

• Design a plain concrete footing to determine if a thinner footing is feasible

• Design thinner, reinforced concrete footing (tradeoff with material and labor).

4. Design a reinforced concrete footing.

Given Square footing, 28 in x 28 in

f'c= 2,500 psi concrete; 60,000 psi steel

Find Footing thickness and reinforcement

b 5 ft 2.2 ft 26 in 2 = = =
Residential Structural Design Guide 4-68

(a) Select trial footing thickness, rebar size, and placement.

t = 6 in

c = 3 in

db = 0.5 in (No. 4 rebar)

(b) Calculate the distance from extreme compression fiber to centroid of reinforcement 

d.

d = t-c-0.5db

= 6 in – 3 in - 0.5 (0.5 in)

= 2.75 in

(c) Determine soil pressure based on factored load.

(d) Check one-way (beam) shear in footing for trial footing thickness.

OK

(e) Check two-way (punching) shear in trial footing.

OK

(f) Determine reinforcement required for footing, based on critical moment at edge of 

column.

OK

Use four No. 4 bars where As = 4(0.2 in2

) = 0.8 in2 ≥ 0.77 in2 OK

Conclusion Use minimum 28-in x 28-in x 6-in footing with four No. 4 bars or three No. 5 bars each 

way in footing.

( )( ) ( )( ) 2,261psf

5 ft

1.2 1,440 lb 300 lb 1.6 5,760 lb

A

P

q 2

footing

u s = + + = =

φVc >> Vu

( ( − )− )= 





 





 = 0.5 b T d

b

P

V u u

φVc >> Vu

φVc > Vu

( ) 2 2

2

u u b (T d)

b

P

V − + 



 





 =

((28 in) (3.5 in 2.75 in) ) 10,741 lbs

(28 in)

11,304 lbs 2 2

2 = − + =
Residential Structural Design Guide 4-69

f'c = 2,500 psi minimum (concrete)

fy = 60,000 psi minimum (steel reinforcing bar)
Residential Structural Design Guide 4-70

EXAMPLE 4.3 Plain Concrete Foundation Wall Design

Given Design loads

Snow load (S) = 280 plf

Live load (L) = 650 plf

Dead load (D) = 450 plf

Moment at top = 0

Concrete weight = 150 pcf

Backfill material = 45 pcf

f'c = 3,000 psi

Wall thickness = 8 in

Wall height = 8 ft

Unbalanced backfill height = 7 ft

Assume axial load is in middle one-third of wall.

Find Verify that an 8-inch-thick plain concrete wall is adequate for the following load 

combinations from chapter 3 (table 3.1).

• 1.2D + 1.6H

• 1.2D + 1.6H + 1.6L + 0.5 (Lr + S)

• 1.2D + 1.6H = 1.6 (Lr + S) + 0.5L

Only the first load combination will be evaluated because it can be shown to govern 

the wall design.

Solution

1. Determine loads.

Equivalent fluid density of backfill soil

Silty clay: w = 100 pcf, Ka = 0.45 (see section 3.5)

q = Kaw = (0.45)(100 pcf) = 45 pcf

Total lateral earth load

Maximum shear occurs at bottom of wall (see figure A.1 of appendix A)

(45pcf)(7ft) 1,103 plf 2

1 ql

2

1 H 2 2 = = =

(45pcf)(7ft) 1,103 plf 2

1 ql

2

1 H 2 2 = = =

(7 ft) 2.33 ft

3

1 l

3

1 X1 = = =

( )( ) 781plf 3(8 ft)

7 ft 45 pcf 7 ft 1

2

1

3L

h qh 1

2

1 V V 2 2

bottom 1 =













  = − 

 



 = = −
Residential Structural Design Guide 4-71

Maximum moment and its location

x = 

= 

= 3.2 ft from base of wall or 4.8 ft from top of wall

Mmax (at x = 3.2 ft) = V1x - qhx2 + qx3

= (781 plf) (3.2 ft) - (45 pcf)(7 ft)(3.2 ft)2 + (45 pcf)(3.2 ft)3

= 1,132 ft-lb/

2. Check shear capacity.

(a) Factored shear load

Vu = 1.6 Vbottom

= 1.6 (781 plf) = 1,250 plf

(b) Factored shear resistance

φ Vn =

(c) Check φVn ≥ Vu

4,557 plf >> 1,250 plf OK

3. Check combined bending and axial load capacity.

(a) Factored loads

Mu = 1.6 Mmax = 1.6 (1,132 ft-lb/lf) = 1,811 ft-lb/lf

Pu = 1.2 D

Dstructure = 450 plf (given)

Dconcrete@x = 

D = 450 plf + 480 plf = 930 plf

Pu = 1.2 (930 plf) = 1,116 plf

(b) Determine Mn, Mmin, Pu

Mn = 0.85 f'cS

S = = 

Mn = 0.85 (3,000 psi)(128 in3

/lf) = 326,400 in-lb/lf = 27,200 ft-lb/lf

Mmin = 0.1hPu = 0.1 (1,112 plf) = 74 ft-lb/lf

Mu > Mmin OK

q

2V h h 2 1 − −

( ) 45 pcf

2(781plf) 7 ft 7 ft 2 − −

f

f bh

3

4 c φ ′

( ) (8 ft 3.23 ft) 480 plf 12 in / ft

8 in 150 plf − = 















2 bd

6

1 (12 in)(8in) 128in / lf

6

1 2 3  = 

 





















12 in / lf

8in
Residential Structural Design Guide 4-72

(c) Check combined bending and axial stress equations

Compression Puu

ΦPnn

+ Muu

ΦMnn

≤ 1

0.11 ≤ 1 OK

Tension

158 ≤ 178 OK

∴ No reinforcement required

4. Check deflection at mid-span (see figure A.1 in appendix A).

=

= 0.009 in/lf

all =

max << all OK

Conclusion An 8-inch-thick plain concrete wall is adequate under the given conditions.

The preceding analysis was performed for a given wall thickness. The same 

equations can be used to solve for the minimum wall thickness that satisfies the 

requirements for shear, combined bending and axial stress, and deflection. With this 

approach to the problem, the minimum thickness would be 7.6 inches (controlled by 

tensile stress under combined bending and axial load). 

1

(0.65)(27,200 ft lb / lf)

1,811ft lb / lf

(0.65)(148,500 plf)

1,116 plf ≤ −

− +

c

g

u u 5 f

A

P

S

M − ≤ φ ′

( ) (0.65)(5) 3,000 psi (8in)(12 in)

1,116 plf

128in / lf

1,811ft lb / lf 12 in / ft

3 − ≤ −

0.4 in / lf

240

(8 ft)(12 in / ft)

240

L = =
Residential Structural Design Guide 4-73

In the strength-based design approach, the safety margin is related to the use of load 

and resistance factors. In this problem, the load factor was 1.6 (for a soil load, H) 

and the resistance factor 0.65 (for tensile bending stress). In terms of a traditional 

safety factor, an equivalent safety margin is found by 1.6/0.65 = 2.5. It is a fairly 

conservative safety margin for residential structures and would allow for an 

equivalent soil fluid density of as much as 113 pcf (45 pcf x 2.5) at the point the 

concrete tensile capacity based on the minimum concrete compressive strength (as 

estimated by ) is realized. This capacity would exceed loads that might be 

expected should the soil become saturated, which would occur under severe 

flooding on a site that is not well drained.

The use of reinforcement varies widely as an optional enhancement in residential 

construction to control cracking and provide some nominal strength benefits. If 

reinforcement is used as a matter of good practice, one No. 4 bar may be placed as 

much as 8 feet on center. One horizontal bar may also be placed horizontally at the 

top of the wall and at mid-height.

c 5 f'
Residential Structural Design Guide 4-74

EXAMPLE 4.4 Reinforced Concrete Foundation Wall Design

Given Service loads

Live load (L) = 1000 plf 

Dead load (D) = 750 plf

Moment at top = 0

Concrete weight = 150 pcf

Backfill material = 60 pcf (equivalent fluid density)

Wall thickness = 8 in

Wall height = 10 ft

Unbalanced backfill height = 8 ft

f'c = 3,000 psi

fy = 60,000 psi

Assume axial load is in middle one-third of wall.

Find If one No. 5 bar at 24 inches on center vertically is adequate for the load 

combination, U = 1.2D + 1.6H + 1.6L (chapter 3, table 3.1) when rebar is placed 3 

inches from outer face of wall (d=5 in).

Solution

1. Determine loads.

Total lateral earth load

Maximum shear occurs at bottom of wall.

∑Mtop = 0

Vbottom =

Maximum moment and its location

Xmax =

=

=

Xmax = 3.87 ft from base of wall or 6.13 ft from top of wall

(60 pcf )(8 ft) 1,920 plf 2

1 ql

2

1 H 2 2 = = =

(8 ft) 2.67 ft

3

1 l

3

1 X = = =

( ) ( )( ) 1,408 plf 10 ft

1,920 plf 10 ft 2.67 ft

L

H L x = − = −

q

ql q l 2qVbottom

2 2 − −

( )

60 pcf

(60 pcf)(8 ft) 60 pcf (8 ft) 2(60 pcf)(1,408 plf) 2 2 − −
Residential Structural Design Guide 4-75

Mmax =

=

= 2,434 ft-lb/lf

2. Check shear capacity, assuming no shear reinforcement is required (Vs=0).

(a) Factored shear load

Vu = 1.6 Vbottom

= 1.6 (1,408 plf) = 2,253 plf

(b) Factored shear resistance

φVn = φ (Vc + Vs)

=

=

(c) Check φVn ≥ Vu

5,587 plf >> 2,253 plf OK

3. Determine slenderness.

All four foundation walls are concrete with few openings; therefore, the system is a 

nonsway frame. This is a standard assumption for residential concrete foundation 

walls.

Slenderness

< 34

∴ Use 

moment magnifier method

4. Determine the magnified moment using the moment magnifier method.

Pu = 1.2D + 1.6L = 1.2 (750 plf) + 1.6 (1,000 plf) = 2,500 plf

Using the approximated moment magnifiers in table 4.4, the moment magnifier 

from the table for a 7.5-inch-thick wall, 10 feet high, is between 1.04 and 1.09. For 

a 9.5-inch-thick wall, the values are between 1 and 1.04.

Through interpolation, δ = 1.04 for a 2,500 plf axial load.

V (x ) 6

qx

2

qlx

bottom max

3

max

2

max + + −

(1,408 plf )(3.87 ft) 6

(60pcf)(3.87 ft)

2

(60 pcf)(8 ft)(3.87 ft)

2 3

+ + −

φ (2) f'

c bwd

(0.85) (2) 3,000 psi (12 in) (5in) = 5,587 plf

( ) 2.31

8in (12 in)

(12 in) (8in) 12

1

A

I

r

3

g

g =





 





= =

r

klu

41.6 34

2.31

(1)(8in)(12 in) = ≥
Residential Structural Design Guide 4-76

5. Check pure bending.

a =

a =

φMn = φAsfy (d- )

= 0.9 (0.155 in2

)(60,000 psi)(5 in- ) 

= 40,577 in-lb/lf = 3,381 ft-lb/lf

φPn = 0

Mu = 2,434 ft-lb/lf from step (1)

 δMu = 1.04 (2,434 ft-lb/lf) = 2,531 ft-lb/lf

By inspection of the interaction diagram, one No. 5 at 24 inches on center is OK 

because δMuPu is contained within the interaction curve. 

6. Check deflection.

max =

= 

= 0.025 in/lf

all =

max << all OK

Conclusion An 8-inch-thick reinforced concrete wall with one vertical No. 5 bar at 24 inches on 

center is adequate for the given loading conditions. 

This analysis was performed for a given wall thickness and reinforcement spacing. 

The same equations can be used to solve for the minimum reinforcement that 

satisfies the requirements for shear, combined bending and axial stress, and 

deflection. This approach would be suitable for a computer spreadsheet design aid. 

A packaged computer software program can also be purchased to perform this 

function; however, certain limitations may prohibit the designer from using design 

recommendations given in this guide.

The use of horizontal reinforcement varies widely as an optional enhancement. If 

horizontal reinforcement is used as a matter of preferred practice to control 

potential cracking, one No. 4 bar placed at the top of the wall and at mid-height 

typically is sufficient.

0.304

0.85 (3,000 psi)(12 in)

(0.155in )(60,000 psi)

0.85 f' b

A f 2

c

s y = =

2

a

2

0.304 in

( ) c g

5 3 3 5 3

E I

36

ql Lx

120L

ql x

36L

ql x

120

q x L l

















+ + − − + −

( )

( )( ) ( )( ) ( )

( )

( )( ) ( )

( )

( )( ) ( )( )

( ) ( )( ) 















































+ −

+ − + −

12

12in 8in 3,122,019psi

36

60pcf 8ft 10ft 6.13ft

120 10ft

60pcf 8ft 6.13ft

36 10ft

60pcf 8ft 6.13ft

120

60pcf 6.13ft 10ft 8ft

ft

1728in 3

5 3

5 3 3

3

3

0.5in / lf

240

(10 ft)(12 in / ft)

240

L = =
Residential Structural Design Guide 4-77

EXAMPLE 4.5 Concrete Lintel

Given f'c = 3,000 psi

fy = 60,000 psi

Dead load = 250 plf

Live load = 735 plf

Span = 6.5 ft

Lintel width = 8 in

Lintel depth = 12 in

Find Minimum reinforcement required

Solution

1. Determine reinforcement required for flexure.

φMn ≥ Mu

Mu = wl

2

12 = 1.2 (250 plf) + 1.6 (735 plf)

12 (6.5 ft)2 = Mu

Mu = 5,197 ft-lb = 62,361 in-lb

φMn = φAsfy (d-0.5a)

d = 12-in depth - 1.5-in cover - 0.375-in stirrup = 10.125 in

a = Asfy

0.85fc

′ b

set Mu = φMn to solve for As

Mu = φAsfy

62,364 in-lb = (0.9) As (60,000 psi) 

0 = 546,750As - 52,941 As

2 – 62,364

As,required = 0.115 in2 

∴ Use one No. 4 bar (As = 0.20 in2

)

Check reinforcement ratio.































 − 0.85 f' b

A f

2

1 d

c

s y

( )( ) 





























 − 0.85 3,000 psi 12 in

A 60,000 psi 10.125 in 0.5 s

0.0025

(10.125in)(8in)

0.2 in

bd

A 2 s ρ = = =

0.021

60,000 psi 87,000

87,000

60,000 psi

0.85(3,000 psi)(0.85)

f 87,000

87,000

f

0.85f

y y

c 1

b =















+ =















+

′β

ρ =
Residential Structural Design Guide 4-78

Because OK

2. Determine shear reinforcement.

φVn ≥ Vu

Vu = = = (6.5 ft) = 4,797 lb

= Span-to-depth ratio, = = = 6.5>5 ∴ Regular beam

φVn = φVc + 0 = = = 7,542 lb

Vu ≤

∴ Stirrups are required

Because φVc > Vu > only the minimum shear reinforcement must be provided.

Av,min = = = 0.034 in2

∴Use No. 3 bars

Shear reinforcement is not needed when > Vu

3,771 lb = 4,797lb - [1.2(250 plf)+1.6(735 plf)]x

x = 0.70ft

Supply No. 3 shear reinforcement spaced 5 in on center for a distance 0.7 ft from the 

supports.

3. Check deflection.

Find x for transformed area

x = 1.95 in

Calculate moment of inertia for cracked section and gross section.

= 0.0012 ρ min

2

wl

2

1.2 (250 plf) + 1.6 (735 plf)

φ 2 f'

c b w d (0.85)(2) 3,000 psi(8in)(10.125in)

3,771 lb 4,797 lb

2

7,542 lb

2

Vc = = < φ

2

φ Vc

y

w

f

50b s

60,000 psi

(50)(8in) 2

(10.125 in)

2

φ Vc

( ) nA (d x) 2 h x x = s −

(0.2 in )(10.125in x) 3,122,019 psi

29,000,000 psi 0.5(8in)(x)

2 2 − 













 =

0 4x 1.86x 18.8 2 = + −

ρmax = 0.75ρb = 0.75(0.021) = 0.016
Residential Structural Design Guide 4-79

Calculate modulus of rupture

Calculate cracking moment

= 10.9 kNm/m

Calculate effective moment of inertia.

Because the cracking moment Mcr is larger than the actual moment Mu, the section is not 

cracked; thus, Ie = Ig.

Calculate deflection

allow = = = 0.33 in

actual =

i(LL) = 5(735 plf)(6.5 ft)4

384(3,122,019 psi)(1,152 in4)(ffff3 1,728 in ⁄ 3) = 0.008 in

i(DL+20%LL) = 

= 0.008 in + 2 (0.0055 in) = 0.02 in

LT << allow OK

Conclusion The minimum reinforcement bar required for an 8-inch x 12-inch concrete lintel 

spanning 6.5 feet is one No. 4 bar.

2 s 3

CR hx nA (d x) 3

1 I = + −

3 2 2 4 (8 in)(1.95 in) (9.29)(0.2 in )(10.125 in 1.95in) 144 in

3

1 = + − =

3 3 4

g (8in)(12 in) 1,152 in

12

1 bh

12

1 I = = =

f 7.5 f' 7.5 3,000 psi 411psi r = c = =

78,912 in lb / lf 6.6 kip ft / lf

(0.5)(12 in)

(411psi)(1,152 in )

Y

f I

M

4

t

g r

cr = = = − = −

240

l ( )( )

240

6.5 ft 12 in / ft

c e

4

384 E I

5 wl

( ( ) ( )( )( ))( )

( )( )( ) 0.006 in

384 3,122,019 psi 1,152 in ft /1,728in

5 250 plf 0.20 735 plf 150 pcf 0.66 ft 1ft 6.5 ft

4 3 3

4

= + +

∆LT = ∆i(LL) + λ∆i(DL+20% LL)
Residential Structural Design Guide 4-80

EXAMPLE 4.6 Unreinforced Masonry Wall Design

Given Live load = 1,300 plf

Dead load = 900 plf

Weight of wall = 52.5 psf

Moment at top = 0

Masonry weight = 120 pcf

Backfill material = 30 pcf

f'm = 1,900 psi

Face shell mortar bedding

Assume axial load is in middle one-third of wall.

Find Verify if a 10-in-thick unreinforced masonry wall is adequate for the ACI 530 load 

combination and 4 ft of unbalanced fill

U = D+H

Solution

1. Determine loads.

Equivalent fluid density of backfill soil (chapter 3)

qs = Kaw = (0.30)(100 pcf) = 30 pcf

Total lateral earth load

R = qsl

2 = (30 pcf)(4 ft)2 = 240 plf

x = = = (4 ft) = 1.33 ft

Maximum shear occurs at bottom of wall

ΣMtop = 0

Vbottom = = =200 plf

2

1

2

1

 3

1

3

1

6L

ql

2

ql 2 3

− 6 (8 ft)

30 pcf (4 ft)

2

30 pcf (4 ft)

2 3

−
Residential Structural Design Guide 4-81

Maximum moment and its location

xm =

xm =

=2.37 ft from base of wall

Mmax =

=

= 204 ft-lb/lf

2. Check perpendicular shear.

= = 1.27>1

Fv =

Fv = 53.3 psi

fv = ( )( )( ) 9.1psi 2 face shells 1.375in 12 in

200 plf 1.5

A

V

2

3

n

=













 =















The shear is assumed to be resisted by two face shells because the wall is 

unreinforced and uncracked.

fv< Fv OK

3. Check axial compression.

An = = (12 in)(2)(1.25 in) = 30 in2

I = bh3 + Ad2

= 529 in4

200 plf (2.37 ft) 6

(30 pcf)(2.37 ft)

2

30 pcf (4 ft)(2.37 ft)

2 3

− + +

Vd

M

200 plf (9.625in)

204 ft − lb / lf (12 in / ft)

( ( )) 













= + − + = +

′ = =

53.3 psi

33in

900 plf 52.5 psf 8 ft 2.37 ft 37 psi 0.45

A

N

37 psi 0.45

120 psi

1.5 f 1.5 1,900 psi 65.4 psi

2 n

v

m

(2b)

12

1

( )

(30 pcf)

30 pcf (4 ft) 30 pcf (4 ft) 2 (30 pcf)(200 plf) 2 2 − −

q

ql q l 2qVbottom

2 2 − −

V (x ) 6

qx

2

qlx

bottom m

3

m m + +
Residential Structural Design Guide 4-82

r =

S =

=

8 ft (12 in 

ft )

4.00 in = 24 < 99

Fa = (0.25 f'm) = (0.25)(1,900 psi) =

= 461 psi

Pmax = FaAn = (461 psi)(33 in2

) = 15,214 plf

P = 900 plf (given for U=D+H)

900 plf < 15,214 plf OK

Check Euler buckling load.

Em = 900f’m = 900 (1,900 psi) = 1.71 x 106 psi

ek = (kern eccentricity)

Pe =

=

= 131,703 plf

P P 0.25Pe OK

Euler buckling loads are calculated by using actual eccentricities from gravity 

loads without including effects of lateral loads.

4. Check combined axial compression and flexural capacity.

M = 204 ft-lb/lf

P = 900 plf

virtual eccentricity e = = = 2.72 in

kern eccentricity ek = = = 3.57 in  GOVERNS

e < ek ∴ Assume section is uncracked

Pe =

r

h

3.57 in

33in

118in

A

S

2

3

n

= =

3

2

m 2

r

e 1 0.577

h

E I 



 



 − π

( )( )

( ) ( )

3

2 2

2 6 4

4.14 in

3.57 in 1 0.577

8 ft 12 in / ft

1.71x10 psi 567 in































 − π

≤

900plf

204 ft − lb / lf (12in / ft)

An

S

2

3

33in

118in

3

2

m 2

r

e 1 0.577

h

E I 



 



 − π

3

2

m

2

r

e 1 0.577

h

E I 



 



 − π
Residential Structural Design Guide 4-83

=

Pe = 131,703 plf

P < 0.25 (131,703 plf) = 32,926 plf OK

fa = = = 27 psi

fb = =

= 29 psi

Fa = 462 psi for h/r ≤ 99

Fb = 0.33 f'm = 0.33 (1,900 psi) = 627 psi

OK

5. Check tension capacity from table 2.2.3.2 for normal to bed joints, hollow, 

ungrouted (type M or S mortar).

Ft ≤ 25 psi

ft =

ft < Ft OK

6. Minimum reinforcement.

Horizontal reinforcement at 24 inches on center vertically

Conclusion An unreinforced masonry wall is adequate for the ACI 530 load combination 

evaluated; however, horizontal reinforcement at 16 inches on center may be 

provided optionally to control potential shrinkage cracking, particularly in long 

walls (i.e., greater than 20 to 30 feet long).

If openings are present, use lintels and reinforcement as suggested in sections 

4.5.2.3 and 4.5.2.4.

Note that the calculations have already been completed and that the maximum 

backfill height calculated for an 8-inch-thick unreinforced masonry wall using 

hollow concrete masonry is about 5 feet, with a safety factor of 4.

3

2

2 4

4.14

3.57 1 0.577

(8 ft (12 in / ft))

(900 plf)(1,900 psi)(567 in )















 



 



 − π

An

P

2 33 in

900 plf

S

M ( )( ) ( )( )

3 118 in

204 ft lb / lf 12 in / ft

8 ft

2.37 ft 900 plf 3.57 in + − 















0.10 1

627 psi

29 psi

462 psi

27 psi + = ≤

1.54 psi

118 in

3,400 ft lb / lf

33 in

900 plf

S

M

A

P

2 3 n

= − − + = − +
Residential Structural Design Guide 4-84

EXAMPLE 4.7 Reinforced Masonry Foundation Wall Design

Given Live load = 1,300 plf

Dead load = 900 plf

Moment at top = 0

Masonry weight = 120 pcf

Wall weight = 52.5 psf

Backfill material = 45 pcf

f'm = 2,000 psi

Face shell mortar bedding

Type M or S mortar

Wall is partially grouted, one core is grouted at 24 inches on center

Assume axial load is in middle one-third of wall

Find Verify if one vertical No. 5 bar at 24 inches on center is adequate for a reinforced 

concrete masonry foundation wall that is 8 feet high with 7 feet of unbalanced 

backfill for the ACI 530 load combination.

U=D+H

Solution

1. Determine loads.

Equivalent fluid density of backfill soil (refer to chapter 3)

q = KaW = (0.45)(100) = 45 pcf

Total lateral earth load

R = ql2 = (45 pcf)(7 ft)2 = 1,103 lb

X = = (7 ft) = 2.33 ft

Maximum shear occurs at bottom of wall.

∑Mtop = 0

Vbottom = =

= 781 plf

Maximum moment and its location

xm =

=

= 3.2 ft from base of wall

2

1

2

1

 3

1

3

1

6L

ql

2

ql 2 3

− 6 (8 ft)

(45 pcf)(7 ft)

2

45 pcf (7 ft)

2 3

−

( )

45 pcf

(45 pcf)(7 ft) 45 pcf (7 ft) 2 (45 pcf)(781plf) 2 2 − −

q

ql q l 2qVbottom

2 2 − −
Residential Structural Design Guide 4-85

Mmax =

=

= 1,132 ft-lb/lf

2. Check perpendicular shear.

= = 1.8 > 1

Fv = 1 ≤ 50 psi

= 1 = 44.7psi < 50 psi

Fv = (44.7 psi)(2-ft grouted core spacing) = 89 psi

Ae = ACMU faceshells + Acore

= (24 in - 8.375 in)(2)(1.375 in) + (1.125 in+1.375 in+5.875 in)(9.625 

in)

= 124 in2

fv = = = 13 psi

fv < Fv OK

This assumes that both mortared face shells are in compression.

3. Check parallel shear.

Foundation walls are constrained against lateral loads by the passive pressure of the 

soil and soil-wall friction. Parallel shear on the foundation wall can be neglected by 

design inspection.

4. Check axial compression.

Ae = 124 in2

I = bh3 + Ad2

= = (8.375 in)(9.625 in - 2(1.375 in))

+ 2 

= 1,138 in4

r = = = 3.03 in

V (x ) 6

qx

2

qlx

bottom m

3

m

2

m + +

(781plf )(3.2 ft) 6

(45 pcf)(3.2 ft)

2

45 pcf (7 ft)(3.2 ft)

2 3

+ + −

Vd

M

(781plf )(9.625in)

1,132 ft − lb / lf (12 in / ft)

f'

m

2,000psi

Ae

V

bd

V = ( ) 2 124in

(781plf)(2ft rebarspacing)

12

1

12

1















 



 



  + − 

 



 2

3

2

1.375 in

2

9.625 in (24 in)(1.375 in) (24 in)(1.375 in) 12

1

Ae

I
Residential Structural Design Guide 4-86

= = 32 < 99

∴Fa=(0.25 f'm)

=0.25 (2,000 psi) = 474 psi

Pmax=FaAe = (474 psi)(124 in2

)=58,776 lb

P = 900 lb

P < Pmax OK

5. Check combined axial compression and flexural capacity.

M = 1,132 ft-lb/lf

P = 900 plf

virtual eccentricity = e =

= =15 in Governs

kern eccentricity = ek =

= =1.9 in

e > ek ∴ Tension on section, assume cracked

fa = = = 14.5 psi

fb = = = 57 psi

fb > fa

∴Assume section is cracked

Fa = 0.25 f'm

r

h

3.03in

8 ft (12 in / ft)





















 



 −

2

140r

h 1

( )( )

( ) 































−

2

140 3.03in

8 ft 12 in / ft 1

P

M

P

M

900plf

1,132 ft − lb / lf(12 in / ft)

Ae

S

Ae

S

( )

2

4

124 in

1,138in 0.5 9.625in

Ae

P

2 124 in

900 plf(2ft)

S

M ( )

3 236.5in

1,132 ft − lb / lf 12 in / ft































 −

2

140r)

h 1 





























 −

2

140r)

h 1
Residential Structural Design Guide 4-87

= 0.25 (2,000 psi) 

= 474 psi

Fb = 0.33 f'm = 0.33 (2,000 psi) = 660 psi

+ ≤ 1

+ = 0.12 ≤ 1 OK

6. Check tension.

Mt = AsdFs

= (0.155 in2)(0.5)(9.625 in)(24,000 psi)

= 17,903 in-lb/lf

M = (1,132 ft-lb/lf)(12 in/ft)

= 13,584 in-lb/lf

M<Mt OK

Conclusion One vertical No. 5 bar at 24 inches on center is adequate for the given loading 

combination. In addition, horizontal truss-type reinforcement is recommended at 

24 inches (that is, every third course of block).

Load combination D+H controls design; therefore, a check of D+L+H is not 

shown.

Table 4.5 would allow a 10-inch-thick solid unit masonry wall without rebar in soil 

with 30 pcf equivalent fluid density. This practice has succeeded in residential 

construction except as reported in places with “heavy” clay soils; therefore, a 

design as shown in this example may be replaced by a design in accordance with 

the applicable residential codes’ prescriptive requirements. The reasons for the 

apparent inconsistency may be attributed to a conservative soil pressure assumption 

or a conservative safety factor in ACI 530 relative to typical residential conditions.















 



 



 −

2

140 (3.03 in)

8 ft(12 in / ft) 1

a

a

F

f

b

b

F

f

474 psi

14.5 psi
Residential Structural Design Guide 4-88

4.10 References

American Concrete Institute (ACI). 2010. Guide to Design of Slabs on Ground.

ACI Standard 360-10. Farmington Hills, MI: ACI. 

———. 2011. Building Code Requirements for Masonry Structures. ACI 

Standard 530. Farmington Hills, MI: ACI.

———. 2011. Building Code Requirements for Structural Concrete and 

Commentary. ACI Standard 318-11. Farmington Hills, MI: ACI.

———. 2010. Residential Code Requirements for Structural Concrete. ACI 

Standard 332-10. Farmington Hills, MI: ACI.

American Forest and Paper Association (AF&PA). 2007. Permanent Wood 

Foundation Design Specification. Washington, DC: AF&PA and 

American Wood Council.

American Society of Civil Engineers (ASCE). 2001. Design Guide for FrostProtected Shallow Foundations, ASCE 32-01. Reston, VA: ASCE.

American Society for Testing & Materials (ASTM). 2011. Standard Penetration 

Test and Split-Barrel Sampling of Soils ASTM Standard D1586-11. West 

Conshohocken, PA: ASTM.

———. 2012a. Standard Specification for Mortar for Unit Masonry. ASTM 

Standard C270-12. West Conshohocken, PA: ASTM.

———. 2012b. Standard Specification for Portland Cement. ASTM Standard 

C150-12. West Conshohocken, PA: ASTM.

———. 2012c. Standard Test Method for Compressive Strength of Cylindrical 

Concrete Specimens. ASTM Standard C39-12. West Conshohocken, PA: 

ASTM.

———. 2012d. Standard Test Method for Slump of Hydraulic Cement Concrete. 

ASTM Standard C143-12. West Conshohocken, PA: ASTM.

———. 2013. Standard Specification for Loadbearing Concrete Masonry Units.

ASTM Standard C90-13. West Conshohocken, PA: ASTM.

American Wood Council (AWC). 2012. National Design Specification for Wood 

Construction. Leesburg, VA: AWC.

American Wood Protection Association. 2013. AWPA Book of Standards. 

Birmingham, AL: AWPA.
Residential Structural Design Guide 4-89

Beall, Christine. 2012. Masonry Design and Detailing for Architects, Engineers, 

and Contractors, 6th ed. New York: McGraw-Hill.

Burnham, T. R. and Johnson, D. 1993. In Situ Foundation Characterization Using 

the Dynamic Cone Penetrometer. Maplewood, MN: Minnesota 

Department of Transportation.

Federal Emergency Management Agency (FEMA). 2009. NEHRP Recommended 

Seismic Provisions for New Buildings and Other Structures, FEMA P1050. Washington, DC: FEMA.

———. 2011a. Coastal Construction Manual, FEMA 55, 3rd ed. Washington, 

DC: FEMA.

———. 2011b. Home Builder’s Guide to Coastal Construction, FEMA P-499. 

Washington, DC: FEMA.

International Code Council (ICC). 2012. International Residential Code. 

Washington, DC: ICC.

National Association of Home Builders (NAHB) Research Center. 1996. Frost 

Protected Shallow Foundations in Residential Construction, 2nd ed. 

Upper Marlboro, MD: NAHB.

National Concrete Masonry Association (NCMA). TEK Notes. Herndon, VA: 

NCMA.

Naval Facilities Engineering Command (NAVFAC). 1986. Foundations and 

Earth Structures, Design Manual 7.02. Washington, DC: U.S. 

Government Printing Office

Portland Cement Association (PCA). 1998. Structural Design of Insulating 

Concrete Form Walls in Residential Construction. Skokie, IL: Portland 

Cement Association.

———. 2007. Prescriptive Design of Exterior Concrete Walls for One- and TwoFamily Dwellings, PCA-100. Skokie, IL: Portland Cement Association.

Pile Buck, Inc. 1990. Coastal Construction. Jupiter, FL: Pile Buck, Inc.

———. 2011. Pile Driving by Pile Buck. Jupiter, FL: Pile Buck, Inc.

Post-Tensioning Institute (PTI). 2008. Design and Construction of PostTensioned Slabs on Ground, DC 10.1-08. Phoenix, AZ: PTI. 
Residential Structural Design Guide 4-90

Roller, John. 1996. Design Criteria for Insulating Concrete Form Wall Systems.

RP, Vol. 116. Prepared by the Construction Technology Lab for the 

Portland Cement Association. Skokie, IL: Portland Cement Association.

The Masonry Society (TMS). 2011. Building Code Requirements for Masonry 

Structures. TMS Standard 402. Longmont, CO: TMS.

U.S. Department of Housing and Urban Development (HUD). 2002. Prescriptive 

Method for Insulating Concrete Forms in Residential Construction, 2d Ed 

Prepared for the U.S. Department of Housing and Urban Development and 

Portland Cement Association by the NAHB Research Center, Inc. 

Washington, DC: HUD.

———. 1998. Testing & Design of Lintels Using Insulating Concrete Forms. 

Prepared for the U.S. Department of Housing and Urban Development and 

the Portland Cement Association by the NAHB Research Center, Inc., 

Upper Marlboro, MD. Skokie, IL: Portland Cement Association.

Westergaard, Harold M. 1926. Analysis of Stresses in Concrete Pavements Due to 

Variations of Temperature. Proceedings, Highway Research Board.

Washington, D.C: DOT.
Residential Structural Design Guide 5-1

CHAPTER 5

Design of Wood 

Framing

5.1 General

This chapter addresses elements of above-grade structural systems in 

residential construction. As discussed in chapter 1, the residential construction 

material most commonly used above grade in the United States is light-frame 

wood; therefore, this chapter focuses on structural design that specifies standard 

dimension lumber and structural wood panels (that is, plywood and oriented 

strand board [OSB] sheathing). Design of the lateral force resisting system (shear 

walls and diaphragms) must be approached from a system design perspective and 

is addressed in chapter 6. Chapter 7 addresses connections; their importance 

relative to the overall performance of wood-framed construction cannot be 

overemphasized. Figure 5.1 shows the basic components and assemblies of a 

conventional wood-frame home; chapter 1 contains more detailed references to 

house framing and related construction topics.

Many elements of a home work together as a system to resist lateral and 

axial forces imposed on the above-grade structure and transfer them to the 

foundation. The above-grade structure also helps resist lateral soil loads on 

foundation walls through connection of floor systems to foundations. The issue of 

system performance, therefore, is most critical in the above-grade assemblies of 

light-frame homes. This chapter addresses system-based design principles within 

the context of simple engineering approaches that are familiar to designers.

The design of the above-grade structure involves (1) floors, (2) walls, and 

(3) roofs.
Residential Structural Design Guide 5-2

FIGURE 5.1 Components and Assemblies of a Conventional WoodFramed Home

Each system can be complex to design as a whole; therefore, simple analysis usually 

focuses on the individual elements (subassemblies) that constitute the system. In some cases, 

“system effects” may be considered in simplified form and applied to the design of certain

elements that constitute specifically defined systems. Structural elements that make up a 

residential structural system include—

• Bending members.

• Columns.

• Combined bending and axial loaded members.

• Sheathing (that is, diaphragm).

• Connections.

EXTERIOR 

LOAD 

ROOF 

SHEATHING

ROOF DECKING

ROOF RAFTER 

(OR TRUSS)

WALL 

SHEATHING

HEADER

JOIST

RIDGE 

MEMBER OR 

BLOCKING

CEILING JOIST (OR 

TRUSS BOTTOM 

CHORD)

FLOOR 

DECKING

INTERIOR NON- LOAD BEARING 

WALL

INTERIOR 

LOAD 

BEARING 

FRAME WALL

BLOCKING (AS 

REQUIRED)

JACK AND 

KING STUDS
Residential Structural Design Guide 5-3

The principal method of design for wood-framed construction has 

historically been allowable stress design (ASD). This chapter uses the most 

current version of the ASD method (AWC, 2012b), although the load and 

resistance factor design (LRFD) method is now available as an alternative (AWC, 

2012b). The ASD method is detailed in the National Design Specification for 

Wood Construction (NDS) and its supplement (NDS-S). The designer is 

encouraged to obtain the NDS commentary to develop a better understanding of 

the rationale and substantiation for the NDS (AWC, 2012a).

This chapter looks at the NDS equations in general and includes design 

examples that detail the appropriate use of the equations for specific structural 

elements or systems in light, wood-framed construction. The discussion focuses 

primarily on framing with traditional dimension lumber but gives some 

consideration to common engineered wood products. Other wood framing 

methods, such as post-and-beam construction, are not explicitly addressed in this 

chapter, although much of the information is relevant. System considerations and 

system factors presented in this chapter are relevant only to light, wood-framed 

construction using dimension lumber, however.

Regardless of the type of structural element to analyze, the designer must 

first determine nominal design loads. The loads acting on a framing member or 

system usually are calculated in accordance with the applicable provisions of the 

locally approved building code and engineering standards. The nominal design 

loads and load combinations used in this chapter follow the recommendations in 

chapter 3 for residential design.

Although prescriptive design tables (that is, span tables) and similar 

design aids commonly used in residential applications are not included herein, the 

designer may save considerable effort by consulting such resources. Most local, 

state, or national model building codes, such as the International Residential 

Code (IRC; ICC, 2012), contain prescriptive design and construction provisions 

for conventional residential construction. Similar prescriptive design aids and 

efficient framing practices can be found in Cost-Effective Home Building: A 

Design and Construction Handbook (NAHBRC, 1994). For high-wind conditions, 

prescriptive guidelines for design and construction can be found in the Wood 

Frame Construction Manual (AWC, 2012c) and ICC 600, Standard for 

Residential Construction in High Wind Regions (ICC, 2008). The designer is also 

encouraged to obtain design data on a variety of proprietary engineered wood 

products that are suitable for many special design needs in residential 

construction. These materials generally should not be viewed as simple “one-toone” substitutes for conventional wood framing, however, and any special design 

and construction requirements should be carefully considered in accordance with 

the manufacturer’s recommendation or applicable code evaluation reports.

5.2 Material Properties

A designer specifying wood materials for residential construction must 

understand the natural characteristics of wood and their effect on the engineering 
Residential Structural Design Guide 5-4

properties of lumber. A brief discussion of the properties of lumber and structural 

wood panels follows.

5.2.1 Lumber

General

As with all materials, the designer must consider wood’s strengths and 

weaknesses. A comprehensive source of technical information on wood 

characteristics is the Wood Handbook: Wood as an Engineering Material, 

Centennial Edition (Forest Products Laboratory, 2013). For the most part, the 

knowledge embodied in the handbook is reflected in the provisions of the NDS 

and NDS-S design data; however, many aspects of wood design require good 

judgment.

Wood is a natural material that, as a structural material, demonstrates 

unique and complex characteristics. Wood’s structural properties can be traced 

back to the material’s natural composition. Foremost, wood is a 

nonhomogeneous, non-isotropic material and thus exhibits different structural 

properties depending on the orientation of stresses relative to the grain of the 

wood. The grain is produced by a tree’s annual growth rings, which determine the 

properties of wood along three orientations: tangential, radial, and longitudinal.

Given that lumber is cut from logs in the longitudinal direction, the grain 

is parallel to the length of a lumber member. Depending on where the lumber is 

cut relative to the center of a log (that is, tangential versus radial), properties vary 

across the width and thickness of an individual member.

Wood Species

Structural lumber can be manufactured from a variety of wood species; 

however, the various species used in a given locality are a function of the 

economy, regional availability, and required strength properties. A wood species 

is classified as either hardwood or softwood. Hardwoods are broad-leafed 

deciduous trees, whereas softwoods (that is, conifers) are trees with needle-like 

leaves and are generally evergreen.

Most structural lumber is manufactured from softwoods because of the 

trees’ faster growth rate, availability, and workability (ease of cutting, nailing, and 

so forth). A wood species is further classified into groups or combinations, as 

defined in the NDS. Species within a group have similar properties and are 

subject to the same grading rules. Douglas-Fir-Larch, Southern Yellow Pine, 

Hem-Fir, and Spruce-Pine-Fir are species groups that are widely used in 

residential applications in the United States.
Residential Structural Design Guide 5-5

Lumber Sizes

Wood members are referred to by nominal sizes (for example, 2x4); 

however, true dimensions are somewhat smaller. The difference occurs during the 

dressing stage of the lumber process, when each surface of the member is planed 

to its final dressed dimension after shrinkage has occurred as a result of the 

drying, or “seasoning,” process. Generally, there is a 1/4- to 3/4-inch difference 

between the nominal and dressed sizes of “dry” sawn lumber. For example, a 2x4 

is about 1.5 inches by 3.5 inches, a 2x10 is 1.5 inches by 9.25 inches, and a 1x4 is 

0.75 inch by 3.5 inches. This guide uses nominal member size, but the designer 

must use the true dimensions of the lumber when analyzing structural 

performance or detailing construction dimensions.

Based on the expected application, the tabulated values in the NDS are 

classified by the species of wood as well as by the nominal size of a member. 

Typical NDS classifications follow.

• Boards are less than 2 inches thick.

• Dimension lumber is 2 to 4 inches thick and a minimum of 2 

inches wide.

• Beams and stringers are a minimum of 5 inches thick, with the 

width at least 2 inches greater than the thickness.

• Posts and timbers are a minimum of 5 inches thick, and the width 

does not exceed the thickness by more than 2 inches.

• Decking is 2 to 4 inches thick and loaded in the weak axis of 

bending for a roof, floor, or wall surface.

Most wood used in light-frame residential construction takes the form of 

dimension lumber.

Lumber Grades

Lumber is graded in accordance with standardized grading rules that 

consider the effect of natural growth characteristics and “defects,” such as knots 

and angle of grain, on the member’s structural properties. Growth characteristics 

reduce the overall strength of the member relative to a “perfect,” clear-grained 

member without any natural defects. Most lumber is visually graded, although it 

may also be machine stress-rated or machine evaluated.

Visually graded lumber is graded by an individual who examines the 

wood member at the mill in accordance with an approved agency’s grading rules. 

The grader separates wood members into the appropriate grade classes. Typical 

visual grading classes in order of decreasing strength properties are Select 

Structural, No. 1, No. 2, and Stud. NDS-S contains more information about grades 

of different species of lumber. The designer should consult a lumber supplier or 

contractor regarding locally available lumber species and grades.

Machine stress-rated (MSR) lumber and machine-evaluated lumber

(MEL) is subjected to nondestructive testing of each piece. The wood member is 
Residential Structural Design Guide 5-6

then marked with the appropriate grade stamp, which includes the allowable 

bending stress (Fb) and the modulus of elasticity (E). This grading method yields 

lumber with more consistent structural properties than results from visual grading 

only.

While grading rules vary among grading agencies, the U.S. Department of 

Commerce has set forth minimums for voluntary adoption by the recognized 

lumber grading agencies. For more information regarding grading rules, refer to 

American Softwood Lumber Voluntary Product Standard (USDOC PS-20), which 

is maintained by the National Institute for Standards and Technology (NIST, 

2010a). The NDS-S lists approved grading agencies and roles.

Moisture Content

Wood properties and dimensions change with moisture content (MC). 

Living wood contains a considerable amount of free and bound water. Free water 

is contained between the wood cells and is the first water to be driven off in the 

drying process. Its loss affects neither volume nor structural properties. Bound 

water is contained within the wood cells and accounts for most of the moisture 

under 30 percent; its loss results in changes in both volume (that is, shrinkage) 

and structural properties. The strength of wood peaks at about 10 to 15 percent 

MC.

Given that wood generally has an MC of more than 30 percent when cut 

and may dry to an equilibrium moisture content (EMC) of 8 to 10 percent in a 

protected environment, it should be sufficiently dried or seasoned before 

installation. Proper drying and storage of lumber minimizes problems associated 

with lumber shrinkage and warping. A minimum recommendation calls for using 

“surface dry” lumber with a maximum 19 percent MC. In uses for which 

shrinkage is critical, specifications may call for “KD-15,” which is kiln-dried 

lumber with a maximum MC of 15 percent. The tabulated design values in the 

NDS are based on an MC of 19 percent for dimension lumber.

The designer should plan for the vertical movement that may occur in a 

structure as a result of shrinkage. For more complicated structural details that call 

for various types of materials and systems, the designer may have to account for 

differential shrinkage by isolating members that will shrink from those that will 

maintain dimensional stability. The designer should also detail the structure such 

that shrinkage is as uniform as possible, thereby minimizing shrinkage effects on 

finish surfaces. When practical, details that minimize the amount of wood 

transferring loads perpendicular to grain are preferable.

Shrink and swell can be estimated in accordance with section 5.3.2 for the 

width and thickness of wood members (that is, tangentially and radially with 

respect to annual rings). Shrinkage in the longitudinal direction of a wood 

member (that is, parallel to grain) is negligible, unless juvenile wood is involved.
Residential Structural Design Guide 5-7

Durability

Moisture is a primary factor affecting the durability of lumber. Fungi, 

which feed on wood cells, require moisture, air, and favorable temperatures to 

survive. When wood is subject to moisture levels above 20 percent and other 

favorable conditions, decay begins to set in; therefore, it is important to protect 

wood materials from moisture by doing the following—

• Limiting end use (for example, by specifying interior applications or 

isolating lumber from ground contact).

• Using a weather barrier (siding, roofing, building wrap, flashing, and 

so on).

• Applying a protective coating (such as paint or water repellent).

• Installing roof overhangs and gutters.

• Specifying preservative-treated or naturally decay-resistant wood.

For homes, an exterior weather barrier (for example, roofing and siding) 

protects most structural wood; however, improper detailing can lead to moisture 

intrusion and decay. Problems are commonly associated with improper or missing 

flashing and undue reliance on caulking to prevent moisture intrusion. For 

additional information and guidance on improving the durability of wood in 

buildings, the designer may refer to Durability by Design (HUD, 2015) and

Moisture Resistant Homes (HUD, 2006).

Wood members that are in ground contact should be preservative treated. 

Preservative treatment is accomplished with a variety of chemicals and chemical 

processes; the process usually is either submersion or pressure injection, and the 

chemicals likely contain a form of copper or borate. Some of the chemicals harm 

metals, so metal connectors used to attach wood to buildings or to connect wood 

members together must be carefully selected. The most common lumber treatment 

historically had been CCA (copper-chromium-arsenate), but this product is 

generally unavailable for use in the residential market. Alternatives treatments for 

wood include alkaline copper quaternary (ACQ), borates and copper azole, in 

addition to other chemical and non-chemical treatments. Designers should 

understand the performance of the wood preservative approach selected.

Termites and other wood-destroying insects (for example, carpenter ants 

and boring beetles) attack wood materials. Some practical solutions include 

chemical treatment of soil, installation of physical barriers (such as termite 

shields), and the use of treated lumber.

Termites are a special problem in warmer climates, although they also 

plague many other areas of the United States. The most common termites are 

“subterranean” termites that nest in the ground and enter wood that is near or in 

contact with damp soil. They gain access to above-grade wood through cracks in 

the foundation or through shelter tubes (that is, mud tunnels) on the surface of 

foundation walls. Because the presence of termites can be visually detected, 

wood-framed homes require periodic inspection for signs of termites.
Residential Structural Design Guide 5-8

Certain wood species are naturally decay resistant, including cedar, 

redwood, cypress, mahogany, and a few others. Some of these products are easy 

to obtain (cedar and redwood), some are expensive (mahogany), and some are in 

short supply (cypress). Even some of those woods are not very decay resistant, 

however, if the wood member is cut from the softest part of the tree. Many 

examples of cedar used outside and placed in contact with the ground reveal that 

decay has occurred in 10 years or less. A designer must exercise care when 

selecting wood to be used outside or in a wet environment.

5.2.2 Wood Structural Panels

Boards historically were used for roof, floor, and wall sheathing; in the 

past 50 years, however, structural wood panel products have come to dominate 

the sheathing market. Structural wood panel products are more economical and 

efficient and typically are stronger than traditional board sheathing. Structural 

wood panel products include plywood and oriented strand board (OSB).

Plywood is manufactured from wood veneers glued together under high 

temperature and pressure. Each veneer or ply is placed with its grain 

perpendicular to the grain of the previous layer. The outer layers are placed with 

their grain parallel to the longer dimension of the panel. Thus, plywood is 

stronger in bending along the long direction and should be placed with the long 

dimension spanning floor and roof framing members. The number of plies 

typically ranges from 3 to 5. OSB is manufactured from thin wood strands glued 

together under high temperature and pressure. The strands are layered and 

oriented to produce properties that are consistent in all directions.

The designer should specify the grade and span rating of structural wood 

panels to meet the required application and loading condition (that is, roof, wall, 

or floor). The most common panel size is 4 feet wide by 8 feet long, with 

thicknesses typically ranging from 1/4 inch to more than 1 inch. Panels can be 

ordered in longer lengths for special applications, especially for tall walls and for 

overlapping sheathing onto floor framing to improve vertical uplift load path 

continuity.

Plywood is performance rated according to the provisions of USDOC PS 

1-09 for industrial and construction plywood (NIST, 2009). OSB products are 

performance rated according to the provisions of USDOC PS 2-10 (NIST, 2010b). 

Those standards are voluntary, however, and not all wood-based panel products 

are rated accordingly. APA—The Engineered Wood Association’s (formerly 

American Plywood Association’s) rating system for structural wood panel 

sheathing products and those used by other structural panel trademarking 

organizations are based on the U.S. Department of Commerce voluntary product 

standards.

The veneer grade of plywood is associated with the veneers used on the 

exposed faces of a panel as follows.

Grade A: The highest quality veneer grade, which is intended for cabinet 

or furniture use.
Residential Structural Design Guide 5-9

Grade B: A high-quality veneer grade, which is intended for cabinet or 

furniture use with all defects repaired.

Grade C: The minimum veneer grade, which is intended for exterior use.

Grade D: The lowest quality veneer grade, which is intended for interior 

use or where protected from exposure to weather.

The wood strands or veneer layers used in wood structural panels are 

bonded with adhesives, and they vary in moisture resistance; therefore, wood 

structural panels are also classified with respect to end-use exposure, as follows.

• Exterior panels are designed for applications with permanent 

exposure to the weather or moisture.

• Exposure 1 panels are designed for applications in which 

temporary exposure to the weather due to construction sequence 

may be expected.

Typical span ratings for structural wood panels specify either the 

maximum allowable center-to-center spacing of supports (for example, 24 inches 

on center for roof, floor, or wall) or two numbers separated by a slash to designate 

the allowable center-to-center spacing of roof and floor supports, respectively (for 

example, 48/24). Even though the second rating method does not specifically 

indicate wall stud spacing, the panels may also be used for wall sheathing. The 

Engineered Wood Construction Guide provides a correlation between roof/floor 

ratings and allowable wall support spacing (APA, 2011a). The Load-Span Tables 

for APA Structural-Use Panels (APA, 2011b) provide span ratings for various 

standard and nonstandard loading conditions and deflection limits.

5.2.3 Lumber Design Values

The NDS-S provides tabulated design stress values for bending, tension 

parallel to grain, shear parallel to grain, compression parallel and perpendicular to 

grain, and modulus of elasticity. The 2012 edition of the NDS (AWC, 2012b) 

includes the design values based on test results conducted prior to 2012. In 2012, 

however, the allowable stress values for southern pine lumber were revised again 

based on very recent testing. Those new values are not reflected in the 2012 NDS, 

although addenda are available with this information and should be used for 

design.

Characteristic structural properties for use in ASD (ASTM D1990) and 

LRFD (ASTM D5457) are used to establish design values (ASTM, 2007; ASTM, 

2012). Test data collected in accordance with the applicable standards determine a 

characteristic strength value for each grade and species of lumber. The value 

usually is the mean (average) or fifth percentile test value. The fifth percentile 

represents the value that 95 percent of the sampled members exceeded. Reduction 

factors are applied to the allowable values published in the NDS-S for 

standardized conditions. The reduction factor normalizes the lumber properties to 

a standard set of conditions related to load duration, MC, and other factors. It also 
Residential Structural Design Guide 5-10

includes a safety adjustment if applicable to the particular limit state (that is, 

ultimate capacity). For specific design conditions that differ from the standard 

basis, therefore, design property values should be adjusted as described in section 

5.2.4.

5.2.4 Adjustment Factors

The allowable values published in the NDS-S are determined for a 

standard set of conditions, yet, many variations in the characteristics of wood 

affect the material’s structural properties. For efficient design, designers must 

make appropriate adjustments for conditions that vary from those used to derive 

the standard design values. Table 5.1 presents adjustment factors that apply to 

different structural properties of wood. The following sections briefly discuss the 

adjustment factors most commonly used in residential applications. For 

information on other adjustment factors, refer to the NDS, NDS-S, and NDS 

commentary.

TABLE 5.1 Adjustment Factor Applicability to Design Values for Wood 

Design Properties1 Adjustment Factor2

CD Cr CH CF CP CL CM Cfu Cb CT CV Ct Ci Cc Cf

Fb X X X X X X X X X X X

Ft X X X X X

Fv X X X X X

X X X X

Fc X X X X X X

E X X X X

1

Basic or unadjusted values for design properties of wood are found in NDS-S. 2

Shaded cells represent factors most commonly used in residential applications; other factors may apply to special conditions.

Source: Based on NDS•2.3 (AWC, 2012b)

Key to Adjustment Factors:

• CD, Load Duration Factor. Applies when loads are other than “normal” 10-year duration (see section 

5.2.4.1 and NDS•2.3.2).

• Cr, Repetitive Member Factor. Applies to bending members in assemblies with multiple members spaced 

at maximum 24 inches on center (see section 5.2.4.2 and NDS•4.3.4).

• CH, Horizontal Shear Factor. Applies to individual or multiple members with regard to horizontal, 

parallel-to-grain splitting (see section 5.2.4.3 and NDS-S).

• CF, Size Factor. Applies to member sizes or grades other than “standard” test specimens, but does not 

apply to Southern Yellow Pine (see section 5.2.4.4 and NDS-S).

• CP, Column Stability Factor. Applies to lateral support condition of compression members (see section 

5.2.4.5 and NDS•3.7.1).

Fc⊥
Residential Structural Design Guide 5-11

• CL, Beam Stability Factor. Applies to bending members not subject to continuous lateral support on the 

compression edge (see section 5.2.4.6 and NDS•3.3.3).

• CM, Wet Service Factor. Applies where the MC is expected to exceed 19 percent for extended periods (see 

NDS-S).

• Cfu, Flat Use Factor. Applies where dimension lumber 2 to 4 inches thick is subject to a bending load in its 

weak axis direction (see NDS-S).

• Cb, Bearing Area Factor. Applies to members with bearing between 3 and 6 inches from the members’ 

ends (see NDS•2.3.10).

• CT, Buckling Stiffness Factor. Applies only to maximum 2x4 dimension lumber in the top chord of wood 

trusses that are subjected to combined flexure and axial compression (see NDS•4.4.3).

• CV, Volume Factor. Applies to glulam bending members loaded perpendicular to the wide face of the 

laminations in strong axis bending (see NDS•5.3.2).

• Ct, Temperature Factor. Applies where temperatures exceed 100o

F for long periods; not normally 

required when wood members are subjected to intermittent higher temperatures, such as in roof structures 

(see NDS•2.4.3 and NDS•appendix C).

• Ci, Incising Factor. Applies where structural sawn lumber is incised to increase penetration of 

preservatives, with small incisions cut parallel to the grain (see NDS•2.3.11).

• Cc, Curvature Factor. Applies only to curved portions of glued laminated bending members (see 

NDS•5.3.4).

• Cf, Form Factor. Applies where bending members are either round or square with diagonal loading (see 

NDS•2.3.8).

5.2.4.1 Load Duration Factor (CD)

Lumber strength is affected by the cumulative duration of maximum 

variable loads experienced during the life of the structure. In other words, strength 

is affected by both the load intensity and its duration (that is, the load history). 

Because of its natural composition, wood is better able to resist higher short-term 

loads (transient live loads or impact loads) than long-term loads (dead loads and 

sustained live loads). Under impact loading, wood can resist about twice as much 

stress as the standard 10-year load duration (that is, “normal duration”) to which 

wood bending stress properties are normalized in the NDS.

When other loads with different duration characteristics are considered, 

certain tabulated stresses must be modified by a load duration factor (CD), as 

shown in table 5.2. Values of CD for various load types are based on the total 

accumulated time effects of a given type of load during the useful life of a 

structure. CD increases with decreasing load duration.

Where more than one load type is specified in a design analysis, the load 

duration factor associated with the shortest duration load is applied to the entire 

combination of loads. For the load combination Dead Load + Snow Load + Wind 

Load, for example, CD is equal to 1.6.
Residential Structural Design Guide 5-12

TABLE 5.2 Recommended Load Duration Factors for ASD

Load Type Load Duration Recommended CD Value

Permanent (dead load) Lifetime 0.9

Normal 10 years 1.0

Occupancy (live load) 10 years to 7 days 1.0 

Snow 1 month to 7 days 1.15 

Temporary construction 7 days 1.25

Wind and seismic1 10 minutes to 1 minute 1.6 

Impact 1 second 2.0

ASD = allowable stress design.

1

The NDS uses a wind and seismic load duration of 10 minutes (CD = 1.6). The factor may be as high as 1.8 for earthquake loads, which 

generally have a duration of less than one minute with a much shorter duration for ground motions in the design-level range.

Source: Based on NDS•2.3.2 and NDS•appendix B (AWC, 2012b)

5.2.4.2 Repetitive Member Factor (Cr)

When three or more parallel dimension lumber members are spaced a 

maximum of 24 inches on center and connected with structural sheathing, they 

constitute a structural “system” with more bending capacity than the sum of the 

single members acting individually. Most elements in a house structure benefit 

from an adjustment for the system strength effects inherent in repetitive members.

The tabulated design values given in the NDS are based on single 

members; thus, an increase in allowable stress is permitted in order to account for 

repetitive members. Although the NDS recommends a repetitive member factor of 

1.15, or a 15-percent increase in bending strength, system assembly tests have 

demonstrated that the NDS repetitive member factor is conservative for certain 

conditions. In fact, test results from several studies support the range of repetitive 

member factors shown in table 5.3 for certain design applications. As shown in 

table 5.1, the adjustment factor applies only to extreme fiber in bending, Fb. Later 

sections of chapter 5 cover other system adjustments related to concentrated 

loads, header framing assemblies, and deflection (stiffness) considerations.
Residential Structural Design Guide 5-13

TABLE 5.3 Recommended Repetitive Member Factors for Dimension 

Lumber Used in Framing Systems1, 2

Application Recommended Cr

Value References

Two adjacent members sharing load3 1.1 to 1.2 AWC, 2012b

HUD, 1999

Three adjacent members sharing load3 1.2 to 1.3 ASAE, 1997

Four or more adjacent members sharing load3 1.3 to 1.4 ASAE, 1997

Three or more members spaced not more than 24 inches on 

center, with suitable surfacing to distribute loads to adjacent 

members (for example, decking, panels, boards,)4

1.15 AWC, 2012a

Wall framing (studs) of three or more members spaced not more 

than 24 inches on center, with minimum 3/8-inch thick wood 

structural panel sheathing on one side and 1/2-inch thick gypsum 

board on the other side5

1.5—2x4 or smaller

1.35—2x6

1.25—2x8

1.2—2x10

AWC, 2012a

Polensek, 1975

1

The National Design Specification for Wood Construction (NDS) recommends a Cr value of 1.15 only as shown in the table. The other 

values in the table were obtained from various codes, standards, and research reports, as indicated. 2

Dimension lumber bending members are to be parallel to each other, continuous (not spliced), and of the same species, grade, and size. The 

applicable sizes of dimension lumber range from 2x4 to 2x12.

3

Cr values are given as a range and are applicable to built-up columns and beams formed of continuous members with the strong axis of all 

members oriented identically. In general, a larger value of Cr should be used for dimension lumber materials that have a greater variability 

in strength (that is, the more variability in strength of individual members, the greater the benefit realized in forming a built-up member 

relative to the individual member strength). A two-ply built-up member of No. 2 grade (visually graded) dimension lumber may qualify for 

use of a Cr value of 1.2, for example, whereas a two-ply member of No. 1 dense or mechanically graded lumber may qualify for a Cr value 

of 1.1. The individual members should be adequately attached to one another or the load introduced to the built-up member so that the 

individual members act as a unit (that is, all members deflect equally) in resisting the bending load. For built-up bending members with 

noncontinuous plies (that is, splices), refer to ASAE EP 559 (ASAE, 1997). For built-up columns subject to weak axis bending load or 

buckling, refer to ASAE EP 559 and NDS•15.3. 4

Refer to NDS•4.3.4 and the NDS Commentary for additional guidance on using the 1.15 repetitive member factor. 5

The Cr values are based on wood structural panel attachment to wall framing using 8d common nails spaced at 12 inches on center. For

fasteners of a smaller diameter, multiply the Cr values by the ratio of the nail diameter to that of an 8d common nail (0.131 inch diameter). 

The reduction factor applied to Cr need not be less than 0.75, and the resulting value of Cr should not be adjusted to less than 1.15. Doubling 

the nailing (that is, decreasing the fastener spacing by one-half) can increase the Cr value by 16 percent (Polensek, 1975). 

With the exception of the 1.15 repetitive member factor, the NDS does not 

recognize the values in table 5.3; therefore, the values in table 5.3 are provided as 

information to the designer for an alternative method based on various sources of 

technical information, including certain standards, code-recognized guidelines, 

and research studies. The designer may pursue the alternative method approach 

using section 104.11 of the IRC (ICC, 2012). For more information on system 

effects, the following references may be helpful.

• “Structural Performance of Light-Frame Truss-Roof Assemblies” (Wolfe, 

1991).

• “Performance of Light-Frame Redundant Assemblies” (Wolfe, 1990).

• “Reliability of Wood Systems Subjected to Stochastic Live Loads” 

(Rosowsky and Ellingwood, 1992).

• “System Effects in Wood Assemblies” (Douglas and Line, 1996).
Residential Structural Design Guide 5-14

• Design Requirements and Bending Properties for Mechanically 

Laminated Columns (EP 559) (ASAE, 1997).

• System Performance of Wood Header Assemblies (HUD, 1999).

• Wall & Floor Systems: Design and Performance of Light-Frame 

Structures (FPRS, 1983).

5.2.4.3 Other Stress Adjustment Factors

Other stress adjustment factors may be applied for a variety of stress 

conditions and design situations. Those adjustment factors consider shear, lumber 

sizes, column stability, beam stability, and others listed in section 5.2.4. The 

designer should refer to the latest version of the NDS (AWC, 2012b) because the 

adjustment factors and design conditions for which they apply are reevaluated 

frequently. The NDS Commentary could be helpful in understanding the 

derivation of the adjustment factors and how they are applied to the structural 

design.

5.3 Structural Evaluation

As with any structural design, the designer should perform several checks 

with respect to various design factors. This section provides an overview of 

checks specified in the NDS and specifies several design concerns that are not 

addressed by the NDS. In general, the two categories of structural design concerns 

are as follows:

Structural Safety (Strength) Structural Serviceability

• Bending and lateral stability

• Horizontal shear

• Bearing

• Combined bending and axial 

loading

• Compression and column 

stability

• Tension

• Connections

• Deflection from bending

• Lateral drift from seismic or wind 

effects

• Floor vibration

• Shrinkage

• Deterioration

The serviceability issues could be described as possible performance 

issues or conditions that must be considered in the design. The body of work 

related to serviceability is not large, thus the designer will need to do some 

research on these issues and apply judgment based on the research to determine 

how to approach designs for serviceability. Clients are increasingly becoming 
Residential Structural Design Guide 5-15

more insistent that their home perform under a variety of load conditions over an 

extended period of time. Issues related to serviceability will be increasingly 

important to the designer. 

The remainder of this chapter applies those design checks to examples of 

different structural systems and elements in a home. In addition, given that the 

intent of this guide is to provide supplemental instruction for the use of the NDS 

in the efficient design of wood-framed homes, the reader is referred to the NDS 

for symbol definitions, as well as other guidance.

5.3.1 Structural Safety Checks

Bending (Flexural) Capacity

The following equations from the NDS determine if a wood member has 

sufficient bending strength. Notches in bending members should be avoided, but 

small notches are permissible; refer to NDS•3.2.3. Similarly, the diameter of 

holes in bending members should not exceed one-third the member’s depth and 

should be located along the center line of the member. Increasing member depth, 

decreasing the clear span or spacing of the member, or selecting a grade and 

species of lumber with a higher allowable bending stress may yield greater 

flexural capacity. Engineered wood products or alternative materials may also be 

considered.

NDS•3.3

basic design check for bending stress

x (applicable adjustment factors, per section 5.2.4)

section modulus of rectangular member

moment of inertia of rectangular member

distance from extreme fiber to neutral axis

Horizontal Shear

Because shear parallel to grain (that is, horizontal shear) is induced by 

bending action, it is also known as bending shear and is greatest at the neutral 

axis. Bending shear is not transverse shear; lumber will always fail in other modes 

before failing in transverse or cross-grain shear because of the longitudinal 

orientation of the wood fibers in structural members.

The horizontal shear force is calculated for solid sawn lumber by 

including the component of all loads (uniform and concentrated) that act 

perpendicular to the bearing surface of the solid member, in accordance with 

b Fb f ≤ ′

Fb = Fb ′

S

M

I

Mc f b = =

6

bd

c

I S

2

= =

12

bd I

3

=

d 2 c = 1

extreme fiber bending stress caused by bending moment 

from transverse load
Residential Structural Design Guide 5-16

NDS•3.4.3. Loads within a distance, d, from the bearing point are not included in 

the horizontal shear calculation; d is the depth of the member for solid rectangular 

members. Transverse shear is not a required design check, although it is used to 

determine the magnitude of horizontal shear by using basic concepts of 

engineering mechanics, as discussed herein.

The following equations from NDS•3.4 for horizontal shear analysis are 

limited to solid flexural members, such as solid sawn lumber, glulam, or 

mechanically laminated beams. Notches in beams can reduce shear capacity and 

should be considered in accordance with NDS•3.4.4. Also, bolted connections 

influence the shear capacity of a beam; refer to NDS•3.4.5. If required, greater 

horizontal shear capacity may be obtained by increasing member depth or width, 

decreasing the clear span or spacing of the member, or selecting another species 

with a higher allowable shear capacity. The general equation for horizontal shear 

stress is discussed in the NDS and in mechanics of materials textbooks. Because 

dimension lumber is solid and rectangular, the simple equation for fv is most 

commonly used.

NDS•3.4

basic design check for horizontal shear

x (applicable adjustment factors per section 5.2.4)

horizontal shear stress (general equation)

for maximum horizontal shear stress at the neutral axis of solid 

rectangular members

Compression Perpendicular to Grain (Bearing)

For bending members bearing on wood or metal, a minimum bearing of 

1.5 inches typically is recommended. For bending members bearing on masonry, 

a minimum bearing of 3 inches typically is advised. The resulting bearing areas 

may not be adequate in the case of heavily loaded members, however. On the 

other hand, they may be too conservative in the case of lightly loaded members. 

The minimum bearing lengths represent good practice.

The following equations from the NDS are based on net bearing area. 

Note that the provisions of the NDS acknowledge that the inner bearing edge 

experiences added pressure as the member bends. As a practical matter, the added 

pressure does not pose a problem because the compressive capacity, F'c⊥, of wood 

increases as the material is compressed. Further, the design value is based on a 

deformation limit, not on failure by crushing. The NDS thus recommends the 

added pressure at bearing edges not be considered. The designer is also alerted to 

the use of the bearing area factor, Cb, which accounts for the ability of wood to 

distribute large stresses originating from a small bearing area not located near the 

end of a member. Examples include interior bearing supports and compressive 

loads on washers in bolted connections.

v Fv f ≤ ′

Fv = Fv ′

Ib

VQ f v =

2A

3V f v =
Residential Structural Design Guide 5-17

NDS•3.10

basic design check for compression perpendicular to grain

x (applicable adjustment factors per section 5.2.4)

stress perpendicular to grain caused by load, P, on net bearing area, 

Ab.

The above equations pertain to bearing that is perpendicular to grain; for 

bearing at an angle to grain, refer to NDS•3.10. The latter condition would apply 

to sloped bending members (such as rafters) notched at an angle for bearing. 

Combined Bending and Axial Loading

Depending on the application and the combination of loads considered, 

some members, such as wall studs and roof truss members, experience bending 

stress in addition to axial loading. The designer should evaluate combined 

bending and axial stresses as appropriate. If additional capacity is required, the 

selection of a higher grade of lumber is not always an efficient solution for 

overstressed compression members under combined axial and bending loads 

because the design may be limited by stability rather than by a stress failure 

mode. Efficiency issues will become evident when the designer calculates the 

components of the combined stress interaction equations that follow and that are 

found in the NDS.

NDS•3.9

Combined bending and axial tension design check

Combined bending and axial compression design check

Compression and Column Stability

For framing members that support axial loads only (that is, columns), the 

designer must consider whether the framing member can withstand the axial 

compressive forces on it without buckling or compressive failure. If additional 

compression strength is required, the designer should increase member size, 

decrease framing member spacing, provide additional lateral support, or select a 

different grade and species of lumber with higher allowable stresses. Improving 

lateral support is usually the most efficient solution when stability controls the 

design (disregarding any architectural limitations). The need for improved lateral 

⊥ ⊥ ≤ ′ c Fc f

⊥ = ⊥ ′ Fc Fc

b

c A

P f ⊥ =

1

F

f

F

f

*

b

b

t

t + ≤ ′

1

F

f f

**

b

b t ≤ −

1

F

f

F

f F 1

f

F

f F 1

f

F

f

2

bE

b1

cE2

c b2

b2

cE1

c b1

b1

2

c

c ≤















 



 



  −



 



 ′ −

+





 



 ′ −

+ 















′
Residential Structural Design Guide 5-18

support will become evident when the designer performs the calculations 

necessary to determine the stability factor, CP, in accordance with NDS•3.7. 

When a column has continuous lateral support in two directions, buckling is not 

an issue, and Cp = 1.0. If, however, the column is free to buckle in one or more 

directions, Cp must be evaluated for each direction of possible buckling. The 

evaluation must also consider the spacing of intermediate bracing, if any, in each 

direction.

NDS•3.7

basic design check for compression parallel to grain

x (applicable adjustment factors from section 5.2.4, including Cp)

fc =

column stability factor

x (same adjustment factors for F'

c except Cp is not used)

KcE of approximately 0.6 may be justified for use with columns that are not 

individual columns but are instead part of a repetitive system of columns (for 

example, studs). Although the NDS is silent on this matter, this approach to 

column design is consistent with the use of Cr for repetitive member stud design. In 

effect, the modulus of elasticity of a system of columns in a repetitive member 

system should not be based on a lower bound modulus of elasticity of a single 

member.

Tension

Relatively few members in light-frame construction resist tension forces 

only. One notable exception occurs in roof framing, in which cross-ties or bottom 

chords in trusses primarily resist tension forces. Other examples include chord 

and collector members in shear walls and horizontal diaphragms, as discussed in 

chapter 6. Another possibility is a member subject to excessive uplift loads, such 

as those produced by extreme wind. In any event, connection design usually is the 

limiting factor in designing the transfer of tension forces in light-frame 

construction (refer to chapter 7). Tension stresses in wood members can be 

checked by using the following equations, in accordance with NDS•3.8.

NDS•3.8

basic design check for tension parallel to grain

c Fc f ≤ ′

Fc = Fc ′

A

P

( )

c

F

F

2c

1 F F

2c

F

F 1

C

*

c

2 cE *

cE c

*

c

cE

p −















 + −

















+

=

2

e

cE

cE

d

K E

F





 





′ =



c * Fc = F

t Ft f ≤ ′

compressive stress parallel to grain caused by axial load, P, acting on 

the member's cross-sectional area, A.
Residential Structural Design Guide 5-19

NDS•3.8

x (applicable adjustment factors per section 5.2.4)

The NDS does not provide explicit methods for evaluating cross-grain 

tension forces and generally recommends the avoidance of cross-grain tension in 

lumber, even though the material is capable of resisting limited cross-grain 

stresses. Design values for cross-grain tension may be approximated using onethird of the unadjusted horizontal shear stress value, Fv.

One application of cross-grain tension in design is in the transfer of 

moderate uplift loads from wind through the band or rim joist of a floor to the 

construction below. If additional cross-grain tension strength is required, the 

designer should increase member size or consider alternative construction details 

that reduce cross-grain tension forces. When excessive tension stress 

perpendicular to grain cannot be avoided, the use of mechanical reinforcement or 

design detailing to reduce the cross-grain tension forces is considered good 

practice (particularly in high-hazard seismic regions) to ensure that brittle failures 

do not occur.

5.3.2 Structural Serviceability

Deflection Caused by Bending

The NDS does not specifically limit deflection but rather defers to 

designer judgment or building code specifications. Nonetheless, with many 

interior and exterior finishes susceptible to damage by large deflections, this guide 

recommends reasonable deflection limits based on design loads for the design of 

specific elements.

The calculation of member deflection is based on the section properties of 

the beam from NDS-S and the member’s modulus of elasticity, with applicable 

adjustments. Generally, a deflection check using the following equations is based 

on the estimated maximum deflection under a specified loading condition. Given 

that wood exhibits time- and load-magnitude-dependent permanent deflection 

(creep), the total long-term deflection can be estimated using terms of two 

components of the load that relate to short- and long-term deflection, based on 

recommendations provided in NDS•3.5.

Ft = Ft ′

A

P ft = stress in tension parallel to grain caused by axial tension load, P, acting 

on the member's cross-sectional area, A
Residential Structural Design Guide 5-20

NDS•3.5

(see table 5.4 for value of denominator)

(see beam equations in appendix A)

If a deflection check proves unacceptable, the designer may increase 

member depth, decrease the clear span or spacing of the member, or select a grade 

and species of wood with a higher modulus of elasticity. Typical denominator 

values used in the deflection equation range from 120 to 600, depending on 

application and designer judgment. Table 5.4 provides recommended deflection 

limits. Certainly, if a modest adjustment to a deflection limit results in a more 

efficient design, the designer should exercise discretion with respect to a possible 

negative consequence, such as vibration or long-term creep. For lateral bending 

loads on walls, a serviceability load for a deflection check may be determined 

from serviceability wind design maps in ASCE 7-10 (ASCE, 2010) and wind 

design pressures calculated using ASCE 7 formulas. Alternatively, ASCE 7-10 

basic wind speeds may be used to determine a 700-year (ultimate) wind load, then 

reduced by a factor of 0.42 to a service-load level.

TABLE 5.4 Recommended Allowable Deflection Limits1

Element or Condition Deflection Limit2 Load Condition

Rafters without attached ceiling finish /180 Lr or S

Rafters with attached ceiling finishes and trusses /240 Lr or S

Ceiling joists with attached finishes /240 Lattic

Roof girders and beams /240 Lr or S

Exterior walls and interior partitions

 With plaster or stucco finish /360 W5 or E

 With other brittle finishes /240 W5 or E

 With interior gypsum board /180 W5 or E

Headers /240 (Lr or S) or L

Floors3 /360 L

Floor girders and beams4 /360 L 1

Values may be adjusted according to designer discretion with respect to potential increases or decreases in serviceability. In some cases, 

a modification may require local approval of a code variance. Some deflection checks may be different or not required, depending on the 

local code requirements. The load condition includes the live or transient load only, not the dead load.

2

 is the clear span in units of inches for deflection calculations. 3

Floor vibration may be controlled by using /360 for spans up to 15 feet and a 1/2-inch limit for spans greater than 15 feet. Wood I-joist 

manufacturers typically recommend /480 as a deflection limit to provide enhanced floor performance and to control nuisance vibrations. 4

Floor vibration may be controlled for combined girder and joist spans of greater than 20 feet by use of a /480 to /600 deflection limit 

for the girder.

5

The wind load is 0.42 times the component and cladding loads determined from ASCE 7-10.

System effects can influence the stiffness of assemblies in a manner 

similar to that of bending capacity (see section 5.3.1). The estimated deflection 

based on an analysis of an element (for example, stud or joist) is multiplied by 

(120 to 600) estimate allow

 ∆ ≤ ∆ =





 



 ∆ ≅

EI

load and span f estimate
Residential Structural Design Guide 5-21

system deflection factors to account for system effect. Typical deflection checks 

on floors under uniform loading can be easily overestimated by 20 percent or 

more. When concentrated loads are considered on typical light-frame floors with 

wood structural panel subflooring, deflections can be overestimated by a factor of 

2.5 to 3 because of the neglect of the load distribution to adjacent framing 

members and partial composite action (Tucker and Fridley, 1999). When 

adhesives attach wood structural panels to wood framing, even greater reductions 

in deflection are realized because of increased composite action (Gillespie, 

Countryman, and Blomquist, 1978; Pellicane and Anthony, 1996). Reductions in 

expected deflections based on system effects do not comply with IRC Table 

R301.7, and the designer should research any deflection reductions to ensure that 

the condition of interest accurately represents the design condition.

Floor Vibration

The NDS does not specifically address floor vibration because it is seen as 

a serviceability rather than a safety issue. In addition, what is considered an 

“acceptable” amount of floor vibration is highly subjective. Accordingly, reliable 

design information on controlling floor vibration to meet a specific level of 

“acceptance” is not readily available. The following rules of thumb are provided 

for the designer who wants to limit vibration beyond that implied by the 

traditional use of a /360 deflection limit (ATC, 1999; Woeste and Dolan, 1998).

• For floor joist spans less than 15 feet, a deflection limit of /360 

considering design live loads only may be used, where  is the clear 

span of the joist in inches.

• For floor joist clear spans greater than 15 feet, the maximum deflection 

should be limited to 0.5 inches.

• For wood I-joists, the manufacturer’s tables that limit deflection to 

/480 should be used for spans greater than 15 feet, where  is the clear 

span of the member in inches.

• When calculating deflection based on the preceding rules of thumb, 

the designer should use a 40 pounds-per-square-foot (psf) live load for 

all rooms—whether or not they are considered sleeping rooms.

• As an additional recommendation, the floor sheathing may be glued 

and mechanically fastened to the floor joists to enhance the floor 

system’s strength and stiffness.

Floor deflections typically are limited to /360 in the span tables published 

in current building codes using a standard deflection check without consideration 

of system effects. For clear spans greater than 15 feet, this deflection limit has 

caused nuisance vibrations that are unacceptable to some building occupants or 

owners. Floor vibration also is aggravated when the floor is supported on a 

bending member (for example a girder) rather than on a rigid bearing wall. It may 

be desirable to design such girders with a smaller deflection limit to control floor 
Residential Structural Design Guide 5-22

vibration, particularly when girder and floor spans have more than a 20-foot total 

combined span (that is, span of girder plus span of supported floor joist).

A frequently occurring nuisance vibration in residential structures is that 

caused by the operation of mechanical equipment, such as washing machines, 

electrical generators, and water circulating pumps. This equipment usually can be 

installed on vibration isolators to help reduce or even eliminate vibration.

For metal-plate-connected wood trusses, strongbacks are effective in 

reducing floor vibration when they are installed through the trusses near the center 

of the span. A strongback is a continuous bracing member, typically a 2x6, 

fastened edgewise to the base of the vertical web of each truss with two 16d nails. 

For longer spans, strongbacks may be spaced at approximately 8-foot intervals 

across the span. Details for strongbacks may be found in the Metal Plate 

Connected Wood Truss Handbook (Callahan, 2004). Alternatively, more stringent 

deflection criteria may be used for the floor truss design.

Shrinkage

The amount of wood shrinkage in a structure depends on the MC of the 

lumber at the time of installation relative to the EMC that the wood will 

ultimately attain in use. Shrinkage is also dependent on the detailing of the 

structure, such as the amount of lumber supporting loads in a perpendicular-tograin orientation (for example, sills, soles, top plates, and joists). MC at 

installation is a function of the specified drying method, jobsite storage practices, 

and climate conditions during construction. Relatively dry lumber (15 percent or 

less MC) minimizes shrinkage problems affecting finish materials and prevents 

loosening or stressing of connections. A less favorable but acceptable alternative 

is to detail the structure in such a way that shrinkage is uniform, dispersed, or 

otherwise designed to minimize problems. This alternative is the de facto choice 

in simple residential buildings.

Shrink and swell across the width or thickness of lumber can be estimated 

by the equation below from ASTM D1990 for typical softwood structural lumber 

(ASTM, 2007). Shrinkage in the longitudinal direction of the member is 

negligible (FPL, 2010).

ASTM D1990•App. X.1

d1 = member width or thickness at moisture content M1

d2 = member width or thickness at moisture content M2

























− −

− −

=

100

a 0.2M

1

100

a 0.2M

1

d d

1

2

2 1
Residential Structural Design Guide 5-23

Lateral Drift

Drift can be both a perception of motion problem and a collapse 

prevention issue from the P-delta effect. Drift can occur from either high winds or 

a seismic event. The problem of motion perception is certainly a serviceability 

issue but is not likely in high winds unless the structure has a height-to-plan ratio 

that would increase the building’s susceptibility to motion. The literature provides 

little guidance on how to determine lateral drift caused by wind. The International 

Building Code (IBC) has lateral drift equations for earthquakes, but, again unless 

the height-to-plan ratio is extremely high, drift usually is not a problem for 

residential structures. Defining the expected performance of the building with 

respect to motion could assist the designer in determining what design issues must 

be considered. For wind, a building with a large height-to-plan ratio and large 

glass areas near the top of the building would suggest the need to limit lateral 

deflection to prevent the glass from cracking and to reduce the perception of 

excessive lateral movement by the building occupants. For earthquake-induced 

motion, drifts also have been shown to correlate well with structural and 

nonstructural (for example, gypsum wall board) damage. Decreasing lateral drift, 

particularly that caused by shear (rather than overturning), is helpful in reducing 

damage following even moderate earthquakes.

Deterioration

Deterioration that affects serviceability is most likely to occur in places 

that would (1) allow water penetration into the building; (2) allow an increase in 

deflection or drift; (3) allow crushing of wood sill plates or wall bottom plates, 

which could create interior cracks in the finishes; or (4) push the exterior wall 

surface out, causing another potential place to trap water or, at a minimum, cause 

an unsightly bulge. Deterioration also can create structural stability problems if 

left unattended.

The designer can help prevent issues with deterioration by using wood that 

will not decay in the presence of moisture and by designing good detailing that 

will help ensure that water management techniques are employed to keep water 

outside the building envelope and away from the interior of the building. Many 

references are available that offer best-practices for minimizing deterioration in 

wood-framed buildings. One such reference that has comprehensive coverage of 

this issue is Moisture Resistant Homes (HUD, 2006).

5.4 Floor Framing

The objectives of good floor system design are—

• To support occupancy live loads and building dead loads adequately.

• To resist lateral forces resulting from wind and seismic loads and to 

transmit the forces to supporting shear walls through the diaphragm 

action provided by the floors.
Residential Structural Design Guide 5-24

• To provide a suitable subsurface for floor finishes.

• To resist vibration and excessive deflection or “bounce.”

• To serve as a thermal barrier over unconditioned areas (for example, 

crawl spaces).

• To serve as a noise barrier from people living below.

• To provide a one- to two-hour fire rating between dwelling units in 

multifamily buildings (refer to local building codes).

5.4.1 General

A wood floor is a horizontal structural system composed primarily of (1) 

joists, (2) beams or girders, and (3) sheathing.

Wood floor systems traditionally have been built of solid sawn lumber for 

floor joists and girders, but parallel chord wood trusses and wood I-joists are 

experiencing substantial market share, offering advantages for dimensional 

consistency and spans. Floor joists are horizontal, repetitive framing members that 

support the floor sheathing and transfer the live and dead floor loads to the walls, 

beams or girders, or columns below. Beams and girders are horizontal members 

that support floor joists not otherwise supported by interior or exterior loadbearing walls. Floor sheathing is a horizontal structural element—usually 

plywood or OSB panels—that directly supports floor loads and distributes them to 

the framing system below. Floor sheathing also provides lateral support to the 

floor joists. As a structural system, the floor provides resistance to lateral building 

loads resulting from wind and seismic forces and thus constitutes a “horizontal 

diaphragm” (refer to chapter 6). See figure 5.2 for an illustration of floor system 

structural elements.
Residential Structural Design Guide 5-25

FIGURE 5.2 Structural Elements of the Floor System

The design approach discussed herein addresses solid sawn lumber floor 

systems in accordance with the procedures specified in the NDS, with appropriate 

modifications as noted. For more information regarding wood I-joists, trusses, and 

other materials, consult the manufacturer’s specifications and applicable code 

evaluation reports.

Section 5.3 discusses the general design equations and design checks for 

the NDS. This section provides detailed design examples that apply the equations 

in section 5.3, tailoring them to the design of elements in a floor system. The next 

sections refer to the span of a member. The NDS defines span as the clear span of 

the member plus one-half the required bearing at each end of the member. This 

guide simply defines span as the distance between the bearing points along the 

center line. This is somewhat easier to determine and is a more conservative 

definition than that of the NDS.

When designing any structural element, the designer must first determine 

the loads acting on the element. Load combinations used in the analysis of floor 
Residential Structural Design Guide 5-26

members in this guide are taken from table 3.1 of chapter 3. Given that only the 

dead loads (D) of the floor system and live loads (L) of occupancy are present in a 

typical floor system, the controlling design load combination for a simply 

supported floor joist is D+L. Although the standard live loads for living and 

sleeping areas have been developed and are in both the building code and ASCE 

7, the designer should explore with the building’s owners how the space is to be 

used and consider any unusual loading as part of the design. Those unusual 

conditions might may include large whirlpool bathtubs, hot tubs, floors used for 

exercise or weightlifting, or floors used for business. For joists with more 

complicated loading, such as cantilevered joists supporting roof framing, the 

following load combinations may be considered in accordance with chapter 3 and 

table 3.4:

D + L

D + 0.75L + 0.75 (Lr or S)

5.4.2 Floor Joist Design

Readily available tables in residential building codes provide maximum 

allowable spans for different species, grades, sizes, and spacing of floor joists; 

therefore, designing conventional floor joists for residential construction is 

usually unnecessary. To obtain greater economy or performance, however, 

designers may wish to create their own span tables or spreadsheets for future use, 

in accordance with the methods shown in this section. Many of the established 

span limits are set by the deflection limits of the building code and not strictly by 

the loads imposed. Joist spans are set using only uniform dead and live loads; no 

concentrated loads are included.

Designers should keep in mind that the grade and species of lumber are 

often a regional choice governed by economics and availability. Some of the most 

common species of lumber for floor joists are Hem-Fir, Spruce-Pine-Fir, DouglasFir, and Southern Yellow Pine. The most common sizes for floor joists are 2x8, 

2x10, and 2x12. Examples 5.1 and 5.2 in section 5.7 illustrate the design of 

typical floor joists, in accordance with the principles discussed earlier.

For different joist applications, such as a continuous multiple span, the 

designer should use the appropriate beam equations (refer to appendix A) to 

estimate the stresses induced by the loads and reactions. Other materials such as 

wood I-joists and parallel chord floor trusses also are commonly used in lightframe residential and commercial construction; refer to the manufacturer’s data 

for span tables for wood I-joists and other engineered wood products. Wood floor 

trusses can be ordered to specification, and they are provided with an engineering 

certification (that is, stamped shop drawings); cold-formed steel floor joists or 

trusses may also be considered. Figure 5.3 illustrates some conventional and 

alternative floor framing members.
Residential Structural Design Guide 5-27

FIGURE 5.3 Conventional and Alternative Floor Framing Members

Notes:

1

Trusses are available with trimmable ends.

2

Cold-formed steel also is used to make floor trusses.

For typical floor systems supporting a concentrated load at or near center 

span, load distribution to adjacent joists can substantially reduce the bending 

stresses or moment experienced by the loaded joist. A currently available design 

methodology may be beneficial for certain applications such as wood-framed 

garage floors that support heavy concentrated wheel loads (Tucker and Fridley, 

1999). Under such conditions, the maximum bending moment experienced by any 

single joist is reduced by more than 60 percent. A similar reduction in the shear 

loading (and end reaction) of the loaded joist also results, with exception for 

“moving” concentrated loads that may be located near the end of the joist, thus 

creating a large transverse shear load with a small bending moment. The 

previously mentioned design methodology for a single, concentrated load applied 

STEEL BAR JOIST

STANDARD DIMENSIONAL LUMBER 

PARALLEL CHORD WOOD TRUSS1

WOOD I-JOIST

COLD-FORMED STEEL JOIST

2
Residential Structural Design Guide 5-28

near mid-span of a repetitive member floor system is essentially equivalent to 

using a Cr factor of 1.5 or more (see section 5.2.4.2). 

Bridging or cross-braces were formerly thought to provide both necessary 

lateral-torsional bracing of dimension lumber floor joists and stiffer floor systems. 

Full-scale testing of 10 different floor systems, however, as well as additional 

testing in completed homes, has conclusively demonstrated that bridging or crossbracing provides negligible benefit to either the load-carrying capacity or stiffness 

of typical residential floors with dimension lumber framing (sizes of 2x6 through 

2x12) and wood structural panel subflooring (NAHB, 1961). Those findings have 

not been proven to apply to other types of floor joists (for example, I-joists and 

steel joists) or for dimension lumber joists greater than 12 inches in depth. 

According to the study, bridging may be considered necessary for 2x10 and 2x12 

dimension lumber joists with clear spans exceeding about 16 feet and 18 feet, 

respectively (based on a 50 psf total design load and l/360 deflection limit). The 

bottom of long span joists has a tendency to buckle slightly or move out of plane. 

Bridging helps to prevent the tendency to buckle, and it helps provide the joist 

with a straight bottom on which to nail drywall or other ceiling materials from the 

space below the floor. The beam stability provisions of NDS•4.4.1 require 

bridging to be spaced at intervals not exceeding 8 feet along the span of 2x10 and 

2x12 joists.

5.4.3 Beam and Girder Design

The decision to use one girder (throughout this guide, the terms beam and 

girder are used interchangeably) over another is a function of cost, availability, 

span and loading conditions, deflection limits, clearance or headroom 

requirements, and ease of construction. Refer to figure 5.4 for illustrations of 

girder types. Girders in residential construction usually are one of the following 

types.

• Built-up dimension lumber.

• Steel I-beam.

• Engineered wood beam (such as laminated veneer lumber, glulam, or 

parallel strand lumber).

• Site-fabricated beam (such as plywood I-beam, plywood box beam and 

flitch plate beam).

• Wood I-joist.

• Metal plate–connected wood truss (such as parallel chord truss).

Built-up beams are constructed by nailing together two or more plies of 

dimension lumber. Because load sharing occurs between the plies (or lumber 

members), the built-up girder is able to resist higher loads than a single member 

of the same overall dimensions. The built-up member can resist higher loads only 

if butt joints are located at or near supports and are staggered in alternate plies. 

Each ply may be face nailed to the previous ply with 10d nails staggered at 12 

inches on center, top to bottom. The design method and equations are the same as 
Residential Structural Design Guide 5-29

those in section 5.4.2 for floor joists; however, the adjustment factors applying to 

design values and loading conditions are somewhat different. The designer needs 

to keep in mind the following.

• Although floor girders are not typically thought of as “repetitive” 

members, a repetitive-member factor is applicable if the floor girder is 

built up from two or more members (three or more, according to the 

NDS).

• The beam stability factor, CL, is determined in accordance with 

NDS•3.3.3; however, for girders supporting floor framing, lateral 

support is considered to be continuous, and CL = 1.

Figure 5.4 illustrates the design of a built-up floor girder.
Residential Structural Design Guide 5-30

FIGURE 5.4 Examples of Beams and Girders 

-
Residential Structural Design Guide 5-31

Steel I-beams often are used in residential construction because of their 

greater spanning capability. Compared with wood members, they span longer 

distances with a shallower depth, primarily because the allowable bending stress 

in a steel beam is greater than that of wood by more than a factor of 10. A 2x4 or 

2x6 usually is attached to the top surface with bolts to provide a fastening surface 

for floor joists and other wood structural members. Although steel beam shapes 

commonly are referred to as I-beams, a typical 8-inch deep, wide flanged Wshaped beam commonly is considered a house beam. Alternatively, built-up coldformed steel beams (that is, back-to-back C-shapes) may be used to construct Ishaped girders. Refer to the Steel Construction Manual (AISC, 2011) and the 

American Iron and Steel Institute’s publication RG-936 for the design of and span 

tables for residential applications of hot-rolled steel sections (AISI, 1993). Span 

tables for structural steel floor beams can also be found in the Beam Series

(NAHBRC, 1981). The Prescriptive Method for Residential Cold-Formed Steel 

Framing (NAHBRC, 1998) and the North American Standard for Cold-Formed 

Steel—Prescriptive Method for One and Two Family Dwellings (AISI, 2012) 

should be consulted for the design of built-up cold-formed steel sections to be 

used as headers and girders.).

Engineered wood beams include I-joists, wood trusses (that is, girder 

trusses), glue-laminated lumber (glulam), laminated veneer lumber, parallel strand 

lumber, and so forth. This guide does not address the design of engineered wood 

girders because product manufacturers typically provide span tables or engineered 

designs that are considered proprietary. Consult the manufacturer for design 

guidelines or completed span tables. The NDS provides a methodology for the 

design of glue-laminated beams (NDS•5), however.

Site-fabricated beams include plywood box beams, plywood I-beams or 

wood I-joist, and flitch plate beams. Plywood box beams are fabricated from 

continuous dimension lumber flanges (typically 2x4 or 2x6) sandwiched between 

two plywood webs; stiffeners are placed at concentrated loads, end bearing points, 

plywood joints, and maximum 24-inch intervals. Plywood I-beams are similar to 

box beams except that the plywood web is sandwiched between dimension 

lumber wood flanges (typically 2x4 or 2x6), and stiffeners are placed at maximum 

24-inch intervals.

Flitch plate beams are fabricated from a steel plate sandwiched between 

two pieces of dimension lumber to form a composite section; thus, a thinner 

member can be used comparison with a built-up wood girder of similar strength. 

The steel plate typically is 1/4- to 1/2-inch thick and about 1/4 inch less in depth 

than the dimension lumber. The sandwich construction usually is assembled with 

through-bolts staggered at about 12 inches on center. Flitch plate beams derive 

their strength and stiffness from the composite action of steel plate and dimension 

lumber joined together. The lumber also provides a medium for fastening other 

materials using nails or screws.

Span tables for plywood I-beams, plywood box beams, steel-wood Ibeams, and flitch plate beams are provided in NAHB’s Beam Series publications 

(NAHBRC, 1981). Refer to the APA’s Product Design Specification (PDS) 

(APA, 1998a) and Supplement (APA, 1998b) for the design method used for 
Residential Structural Design Guide 5-32

plywood box beams. Section R602.7.2 of the IRC (ICC, 2012), provides a simple 

prescriptive table for plywood box beam headers.

5.4.4 Subfloor Design

Typical subfloor sheathing is nominal 5/8- or 3/4-inch thick 4x8 panels of 

plywood or OSB, with tongue-and-groove edges at unsupported joints 

perpendicular to the floor framing. Sheathing products generally are categorized 

as wood structural panels and are specified in accordance with the prescriptive 

span rating tables published in a building code or are made available by the 

manufacturer. The prescriptive tables provide maximum spans (joist spacing) 

based on sheathing thickness and span rating, with the standard beam calculation 

as the basis. If loads exceed the limits shown in the prescriptive tables, the 

designer may have to perform calculations; however, such calculations are rarely 

necessary.

The APA also recommends a fastener schedule for connecting sheathing 

to floor joists. Generally, nails are placed a minimum of 6 inches on center at 

edges and 12 inches on center along intermediate supports. Table 5.5 lists 

recommended nail sizes based on sheathing thickness. Nail sizes vary with nail 

type (for example, sinkers, box nails, and common nails), and various nail types 

have different characteristics that affect structural properties (refer to chapter 7). 

For information on other types of fasteners, consult the fastener manufacturer. In 

some cases, shear loads in the floor diaphragm resulting from lateral loads (that is, 

wind and earthquake) may require a more stringent fastening schedule; chapter 6 

includes a discussion on fastening schedules for lateral load design. Regardless of 

fastener type, gluing the floor sheathing to the joists increases floor stiffness and 

strength. Gluing also reduces or eliminates floor squeaks, which result when the 

unsecured floor sheathing rubs across the top of the floor joist when someone 

steps on the sheathing. 

TABLE 5.5 Fastening Floor Sheathing to Structural Members1

Thickness Size and Type of Fastener

Plywood and wood structural panels, subfloor sheathing to framing

1/2 inch and less 6d common nail

19/32 to 1 inch 8d common nail

1-1/8 to 1-1/4 inch 10d common nail or 8d deformed shank nail

Plywood and wood structural panels, combination subfloor/underlayment to framing

3/4 inch and less 8d common nail or 6d deformed shank nail

7/8 to 1-inch 8d common nail

1-1/8 to 1-1/4 inch 10d common nail or 8d deformed shank nail

Notes:

1

Codes generally require common or box nails; if pneumatic nails are used, as is common, refer to ESR-1539 (ICC-ES, 2013) or the nail 

manufacturer’s data. Screws also are commonly substituted for nails. For more detail on fasteners and connections, refer to chapter 7.
Residential Structural Design Guide 5-33

5.5 Wall Framing

The objectives of wall system design are—

• To resist snow, live and dead loads, and wind and seismic forces.

• To provide an adequate subsurface for wall finishes and to provide 

openings for doors and windows.

• To serve as a thermal and weather barrier.

• To provide space for and access to electrical and mechanical 

equipment, where required.

• To provide a 1- to 2-hour fire barrier if the wall separates 

individual dwelling units in attached or multifamily buildings.

5.5.1 General

A wall is a vertical structural system that supports gravity loads from the 

roof and floors above and transfers the loads to the foundation below. It also 

resists lateral loads resulting from wind and earthquakes. A typical wood-framed 

wall is composed of the following elements, as shown in figure 5.5.

• Studs, including wall, cripple, jack, and king studs.

• Top and bottom (sole) plates.

• Headers across openings.

• Sheathing.

• Diagonal let-in braces, if used.

Residential wall systems traditionally have been constructed of dimension 

lumber, usually 2x4 or 2x6, although engineered wood studs and cold-formed 

steel studs are seeing increased use. Wall studs are vertical, repetitive framing 

members spaced at regular intervals to support the wall sheathing. They span the 

full height of each story and support the building loads above. King and jack studs 

(also known as jamb studs) frame openings and support loads from a header. 

Cripple studs are placed above or below a wall opening and are not full height. 

Built-up wall studs that are assembled on the jobsite may be used within the wall 

to support concentrated loads. Top and bottom plates are horizontal members to 

which studs are fastened. The top and bottom plates are then fastened to the floor 

or roof above and either to the floor below or directly to the foundation. Headers 

are beams that transfer the loads above an opening to jack studs at each side of the 

opening. Diagonal let-in bracing can be an alternative to wall sheathing to provide 

shear resistance for the wall.
Residential Structural Design Guide 5-34

FIGURE 5.5 Structural Elements of the Wall System

Structural wall sheathing, such as plywood or OSB, distributes lateral 

loads to the wall framing and provides lateral support to both the wall studs (that 

is, buckling resistance) and the entire building (that is, racking resistance). Interior 
Residential Structural Design Guide 5-35

wall finishes also provide significant support to the wall studs and the structure. In 

low-wind and low-hazard seismic areas, metal T braces or wood let-in braces may 

be used in place of wall sheathing to provide resistance to lateral (racking) loads. 

About 80 percent of new homes constructed each year now use wood structural 

panel bracing, and many of those homes are fully sheathed with wood structural 

panels. Chapter 6 addresses the design of wall bracing in greater detail. 

The design procedure discussed herein addresses dimension lumber wall 

systems according to the NDS. Where appropriate, modifications to the NDS have 

been incorporated and are noted. Standard design equations and design checks for 

the NDS procedure were presented earlier in this chapter. The detailed design 

examples in this section illustrate the application of the equations by tailoring 

them to the design of the elements that make up residential wall systems.

Wall systems are designed to withstand dead and live gravity loads acting 

parallel to the wall stud length, as well as lateral loads–primarily wind and 

earthquake loads–acting perpendicular to and parallel to the face of the wall. 

Wind also induces uplift loads on the roof; when the wind load is sufficient to 

offset dead loads, walls and internal connections must be designed to resist 

tension or uplift forces. The outcome of the design of wall elements depends on 

the degree to which the designer uses the “system strength” inherent in the 

construction. To the extent possible, guidance on system design in this section 

uses the NDS and the recommendations in sections 5.2 and 5.3.

When designing wall elements, the designer needs to consider the load 

combinations discussed in chapter 3, particularly the following ASD 

combinations of dead, live, snow, seismic, and wind loads—

• D + 0.75L + 0.75(Lr or S).

• D + (Lr or S). 

• D + 0.75L +0.45W +0.75(Lr or S or R).

• D + 0.52E + 0.75L + 0.75S.

A wall system may support a roof only or a roof and one or more stories 

above. The roof may or may not include an attic storage live load. A 10 psf attic 

live load used for the design of ceiling joists is intended primarily to provide safe 

access to the attic, not support for attic storage. The controlling load combination 

for a wall that supports only a roof is the second load combination listed above. 

The controlling load combination for a wall that supports a floor, wall, and a roof 

should be either the first or second load combination depending on the relative 

magnitude of floor and roof snow loads.

The third load combination provides a check for the out-of-plane bending 

condition due to lateral wind loads on the wall. For tall, wood-framed walls that 

support heavy claddings such as brick veneer, the designer should also consider 

out-of-plane bending loads resulting from an earthquake load combination, 

although the other load combinations shown usually control the design. The third 

and fourth load combinations are essentially combined bending and axial loads 

that may govern stud design, as opposed to axial load only, considered in the first 

two load combinations. Chapter 6 addresses the design of walls for in-plane shear 
Residential Structural Design Guide 5-36

or racking forces resulting from lateral building loads caused by wind or 

earthquakes.

In many cases, certain design load combinations or load components can 

be dismissed or eliminated through practical consideration and inspection. They 

are a matter of designer judgment, experience, and knowledge of the critical 

design conditions.

5.5.2 Load-Bearing Walls

Exterior load-bearing walls support both axial and lateral loads. For 

interior load-bearing walls, only gravity loads are considered. A serviceability 

check using a lateral load of 5 psf is sometimes applied independently to interior 

walls but should not normally control the design of load-bearing framing. This 

section focuses on the axial and lateral load-bearing capacity of exterior and 

interior walls.

Exterior walls are not necessarily always load-bearing walls. Load-bearing 

walls support gravity loads from the roof, ceiling, or floor joists or the beams 

above. A gable-end wall typically is considered a non-load-bearing wall because 

roof and floor framing generally runs parallel to the gable end; however, a gableend wall must support lateral wind and seismic loads and even small dead and live 

loads. Exterior load-bearing walls must be designed for axial loads as well as for 

lateral loads from wind or seismic forces. They must also act as shear walls to 

resist racking loads from lateral wind or seismic forces on the overall building 

(refer to chapter 6). Example 5.4 demonstrates the design of an exterior loadbearing wall.

When calculating the column stability factor for a stud wall, column 

capacity is determined by using the slenderness ratio about the strong axis of the 

stud (le/d)x in accordance with NDS•3.7.1. The strong axis slenderness ratio is 

used because the wall sheathing and finish materials in the stud’s weak-axis 

bending or buckling direction provide lateral support to the stud. When 

determining the column stability factor, Cp, for a wall system rather than for a 

single column, in accordance with NDS•3.7.1, the designer must exercise 

judgment with respect to the calculation of the effective length, e, and the depth 

or thickness of the wall system, d. A buckling coefficient, Ke, of about 0.8 is 

reasonable (see appendix G of NDS) and is supported in the research literature on 

this topic for sheathed wall assemblies and studs with square-cut ends (that is, not 

a pinned joint).

In cases in which continuous support is not present (such as during 

construction), the designer may want to consider stability for both axes. 

Unsupported studs generally fail because of weak-axis buckling under a 

significantly lighter load than would otherwise be possible with continuous lateral 

support in the weak-axis buckling direction, which is provided by sheathing.

Interior walls may be either load bearing or non-load bearing. Non-loadbearing interior walls often are called partitions. In either case, interior walls 

should be solidly fastened to the floor and ceiling framing and to the exterior wall 

framing where they adjoin. Installing extra studs, blocking, or nailers in the 
Residential Structural Design Guide 5-37

outside walls may be necessary to provide for attachment of interior walls. The 

framing also must be arranged to provide a nailing surface for wall-covering 

materials at inside corners. Interior load-bearing walls typically support the floor 

or ceiling joists above when the clear span from exterior wall to exterior wall is 

greater than the spanning capability of the floor or ceiling joists. Interior walls, 

unlike exterior walls, seldom experience large transverse (that is, out of plane) 

lateral loads; however, some building codes require interior walls to be designed 

for a minimum lateral load, such as 5 psf, for serviceability. Some interior loadbearing walls may be shear walls to assist in transferring lateral loads to lower 

floors and to the foundation. If the interior wall is required only to resist axial 

loads, the designer may follow the design procedure demonstrated in example 5.4 

for the axial-load-only case. Generally, axial load design provides more than 

adequate resistance to a nominal lateral load.

If local code requirements require wall studs to be designed to withstand a 

minimum lateral load, the designer should specify load-bearing walls in 

accordance with the previous section on exterior load-bearing walls. (Note that 

the load duration factor, CD, of 1.6 is used for allowable stress increases for 

exterior load-bearing walls when wind or earthquake loads are considered, 

whereas a load duration factor of 1.0 to 1.25 may be used for interior load-bearing 

walls and exterior walls analyzed for live and snow loads; refer to section 

5.2.4.1.)

5.5.3 Headers

Load-bearing headers are horizontal members that carry loads from a wall, 

ceiling, or floor or roof above and transfer the combined load to jack and king 

studs on each side of a window or door opening. The span of the header may be 

taken as the width of the rough opening measured between the jack studs 

supporting the ends of the header. Headers usually are built up from two nominal 

2-inch thick members or are engineered wood beams.

Load-bearing header design and fabrication is similar to that for girders 

(see section 5.4.3). The minimum allowable deflection limit of /240 will permit 

substantial deflection in large openings, which could create a binding on windows 

and doors. For a 10–foot wide opening, for example, the permissible deflection is 

1/2 inch. Large openings or especially heavy loads may require stronger 

members, such as engineered wood beams, hot-rolled steel, or flitch plate beams.

Headers generally are designed to support all loads from above; however, 

typical residential construction calls for a double top plate above the header. 

When an upper story is supported, a floor band joist and sole plate of the wall 

above also are spanning the wall opening below. These elements are all part of the 

load-resisting system. Recent header testing was done to determine whether an 

adjustment factor (that is, a system factor or repetitive-member factor) is justified 

in designing a header (HUD, 1999). The results showed that a repetitive-member 

factor is valid for headers constructed of only two members, as shown in table 

5.3, and that additional system effects produce large increases in capacity when 

the header is overlaid by a double top plate, band joist, and sole plate, as shown in 
Residential Structural Design Guide 5-38

example 5.5. Consequently, an overall system factor of 1.8 was found to be a 

simple, conservative design solution. That system factor is applicable to the 

adjusted bending stress value, Fb', of the header member only. Although this 

example covers only a very specific condition, it exemplifies the magnitude of 

potential system effect in similar conditions. In this case, the system effect is 

associated with load sharing and partial composite action.

Headers are not required in non-load-bearing walls. Openings can be 

framed with single studs and a horizontal header block of the same size. It is 

common practice to use a double 2x4 or triple 2x4 header for larger openings in 

non-load-bearing walls. In the interest of added rigidity and fastening surface, 

however, some builders use additional jamb studs for openings in non-loadbearing walls, but such studs are not required.

5.5.4 Columns

Columns are vertical members placed where an axial force is applied 

parallel to the longitudinal axis. Columns may fail by either crushing or buckling. 

Longer columns have a higher tendency than shorter columns to fail as a result of 

buckling. The load at which the column buckles (Euler buckling load) is directly 

related to the ratio of the column’s unsupported length to its depth (slenderness 

factor). The equations provided in section 5.3 are based on the NDS•3.7.1 

provisions regarding the compression and stability of an axial compression 

member (that is, column) and thus account for the slenderness factor. Example 5.6 

demonstrates column design.

Figure 5.6 illustrates three ways to construct columns using lumber. 

Simple columns are columns fabricated from a single piece of sawn lumber; 

spaced columns are fabricated from two or more individual members with their 

longitudinal axes parallel and separated with blocking at their ends and 

midpoint(s); built-up columns are solid columns fabricated from several 

individual members fastened together. Spaced columns as described in the NDS 

are not normally used in residential buildings and are not addressed here (refer to 

NDS•15.2 for the design of spaced columns).

Steel jack posts also are commonly used in residential construction; 

however, jack post manufacturers typically provide a rated capacity so that no 

design is required except the specification of the design load requirements and the 

selection of a suitable jack post that meets or exceeds the required loading. 

Typical 8-foot tall steel jack posts are made of pipe and have adjustable bases for 

floor leveling. The rated (design) capacity generally ranges from 10,000 to 20,000 

pounds, depending on the steel pipe diameter and wall thickness.

Simple columns are fabricated from one piece of sawn lumber. In 

residential construction, simple columns such as a 4x4 are common. The 

equations in section 5.3 are used to design simple columns, as demonstrated in 

example 5.6.

Built-up columns are fabricated from several wood members fastened 

together with nails or bolts. They are commonly used in residential construction 
Residential Structural Design Guide 5-39

because smaller members can be easily fastened together at the jobsite to form a 

larger column with adequate capacity. 

The nails or bolts used to connect the separate members of a built-up 

column do not rigidly transfer shear loads; therefore, the bending load capacity of 

a built-up column is less than a single column of the same species, grade, and 

cross-sectional area when bending direction is perpendicular to the laminations 

(that is, all members bending in their individual weak-axis direction). The 

coefficient, Kf, accounts for the capacity reduction in bending load for nailed or 

bolted built-up columns. Kf applies, however, only to the weak-axis buckling or 

bending direction of the individual members and therefore should not be used to 

determine Cp for column buckling in the strong-axis direction of the individual 

members. (Refer to NDS•15.3 for nailing and bolting requirements for built-up 

columns.)

The preceding consideration is not an issue when the built-up column is 

sufficiently braced in the weak-axis direction (that is, embedded in a sheathed 

wall assembly). In this typical condition, the built-up column actually is stronger 

than a solid sawn member of equivalent size and grade because of the repetitivemember effect on bending capacity (see table 5.3). When the members in the 

built-up column are staggered or spliced, however, the column bending strength is 

reduced. Although the NDS•15.3 provisions apply only to built-up columns with 

all members extending the full height of the column, design methods for spliced 

columns are available (ASAE, 1997).

FIGURE 5.6 Wood Column Types

SIMPLE COLUMN SPACED COLUMN BUILT-UP COLUMN (WITH 

CONTINUOUS MEMBERS)
Residential Structural Design Guide 5-40

5.6 Roofs

The objectives of roof framing design are—

• To support building dead and snow loads and to resist wind and 

seismic forces.

• To resist roof construction and maintenance loads.

• To provide a thermal and weather barrier.

• To provide support for interior finishes.

• To provide attic space and access for electrical and mechanical 

equipment or storage.

5.6.1 General

A roof in residential construction typically is a sloped structural system 

that supports gravity and lateral loads and transfers the loads to the walls below. 

Generally, the four options for wood roof construction are—

• Roof trusses.

• Rafters and ceiling joists (with or without rafter or collar ties).

• Rafters with ridge beams (that is, cathedral ceilings).

• Timber framing.

By far the most common types of residential roof construction use lightframe trusses, rafters, or a mix of those materials, depending on roof layout. 

Figure 5.7 depicts conventional roof construction and roof framing elements. 

Rafters are repetitive framing members that support the roof sheathing and 

typically span from the exterior walls to a nonstructural ridge board. Rafter pairs 

may also be joined at the ridge with a gusset, thereby eliminating the need for a 

ridge board. Rafters may also be braced at or near mid-span using intermittent 2x 

vertical braces and a 2x runner crossing the bottom edges of the rafters. Ceiling 

joists are repetitive framing members that support ceiling and attic loads and 

transfer the loads to the walls and beams below. They are not typically designed 

to span between exterior walls and therefore require an intermediate bearing wall. 

Overhangs are framed extensions of the roof that protrude beyond the exterior 

wall of the home, typically by 1 to 2 feet. Overhangs protect walls and windows 

from direct sun and rain and therefore offer durability and energy efficiency 

benefits.

Ceiling joists typically are connected to rafter pairs to resist outward thrust 

generated by loading on the roof. Where ceiling joists or collar ties are eliminated 

to create a cathedral ceiling, a structural ridge beam must be used to support the 

roof at the ridge and to prevent outward thrust of the bearing walls. Ceiling joists 

and roof rafters are bending members that are designed similarly; therefore, this 

chapter groups them under one section.
Residential Structural Design Guide 5-41

FIGURE 5.7 Structural Elements of a Conventional Roof System

Roof trusses are pre-engineered components. They are fabricated from 2-

inch thick dimension lumber connected with metal truss plates. They are generally 

more efficient than stick framing and usually are designed to span from exterior 

wall to exterior wall with no intermediate support. Complex portions of roof 

systems, however, often use rafter-framing techniques.

Roof sheathing is a thin structural element, usually plywood or OSB, 

which supports roof loads and distributes lateral and axial loads to the roof 

framing system. Roof sheathing also provides lateral support to the roof framing 

members and serves as a membrane or diaphragm to resist and distribute lateral 

building loads from wind or earthquakes (refer to chapter 6).

Roof systems are designed to withstand dead, live, snow, and wind uplift 

loads; in addition, they are designed to withstand lateral loads, such as wind and 

earthquake loads, transverse to the roof system. The design procedure discussed 

herein addresses dimension lumber roof systems designed according to the NDS. 

Where appropriate, the procedure incorporates modifications of the NDS. Section 

5.3 summarizes the general design equations and design checks based on the 

NDS. Chapter 6 addresses the design of roofs with respect to lateral loads on the 

overall structure, and chapter 7 provides guidance on the design of connections.

When designing roof elements or components, the designer needs to 

consider the following load combinations from chapter 3 (table 3.1).
Residential Structural Design Guide 5-42

• D + (Lr or S or R).

• 0.6 D + 0.6Wu.

• D + 0.6W.

The following sections refer to the span of the member. As mentioned 

previously, the NDS defines span as the clear span of the member plus one-half 

the required bearing at each end of the member. For simplicity, the clear span

between bearing points is used herein.

Finally, roofs exhibit system behavior that is in many respects similar to 

floor framing (see section 5.4); however, sloped roofs also exhibit unique system 

behavior. For example, the sheathing membrane or diaphragm on a sloped roof 

acts as a folded plate that helps resist gravity loads. The effect of the folded plate 

becomes more pronounced as roof pitch becomes steeper. Such a system effect 

usually is not considered in design but explains why light wood-framed roof 

systems may resist loads several times greater than their design capacity. Recent 

research on trussed roof assemblies with wood structural panel sheathing points to 

a system capacity increase factor of 1.1 to 1.5 relative to the design of an 

individual truss (Mtenga, 1998; Wolfe, 1996; Wolfe and LaBissoniere, 1991). 

Again, this system capacity increase factor is not recognized by either the NDS or 

the IRC (ICC, 2012), but it could potentially be used as an alternate method.

5.6.2 Conventional Roof Framing

This section addresses the design of conventional roof rafters, ceiling 

joists (collar ties), ridge beams, and hip and valley rafters. The design procedure 

for a rafter and ceiling joist system is similar to that for a truss except that the 

components and connections are site built. Standard pin-joint analysis is common 

practice to determine axial forces in the members and shear forces at their 

connections. The ceiling joists and rafters are then usually sized according to their 

individual applied bending loads, taking into account axial load effects on the 

members as appropriate. Frequently, intermediate rafter braces (rafter ties) that 

are similar to truss web members are also used when ceiling joists are not 

connected to or are not parallel with the rafters. Standard construction details and 

span tables for rafters and ceiling joists can be found in the IRC (ICC, 2012). 

These tables generally provide allowable horizontal rafter spans, disregarding any 

difference that roof slope may have on axial and bending loads experienced in the 

rafters. This approach generally is considered standard practice. Example 5.7 

demonstrates two design approaches for a simply supported, sloped rafter, as 

illustrated in figure 5.8.

Structural ridge beams are designed to support roof rafters at the ridge 

when no ceiling joists or collar ties are used to resist the outward thrust of rafters. 

A repetitive-member factor, Cr, is applicable if the ridge beam is composed of two 

or more members (see table 5.4). Any additional roof system benefit, such as the 

folded plate action of the roof-sheathing diaphragm, goes ignored in its structural 
Residential Structural Design Guide 5-43

contribution to the ridge beam, particularly for steep-sloped roofs. Example 5.8 

demonstrates the design approach for ridge beams.

Roofs with hips and valleys are constructed with rafters framed into a hip 

or valley rafter as appropriate and, in practice, typically are one to two sizes larger 

than the rafters they support (for example, 2x8 or 2x10 hip for 2x6 rafters). While 

hip and valley rafters experience a unique tributary load pattern or area, they 

generally are designed much like ridge beams. The folded plate effect of the roofsheathing diaphragm provides support to a hip or valley rafter in a manner similar 

to that discussed for ridge beams. Beneficial system effect goes ignored because 

of the lack of definitive technical guidance. 

Many residential buildings are designed with cathedral ceilings. The 

framing method used for those high ceilings does not consider the ceiling as a 

brace at the top of the exterior walls because it is not in a plane that is 

perpendicular to the walls. A hinge forms in the framing at the cathedral ceiling–

exterior wall connection on the gable end if the wall studs do not extend all the 

way to the roofline.

5.6.3 Roof Trusses

Roof trusses incorporate rafters (top chords) and ceiling joists (bottom 

chords) into a structural frame fabricated from 2-inch thick dimension lumber, 

usually 2x4 or 2x6. A combination of web members are positioned between the 

top and bottom chords, usually in triangular arrangements that form a rigid 

framework. Many different truss configurations are possible, including open 

trusses for attic rooms and cathedral or scissor trusses with sloped top and bottom 

chords. The wood truss members are connected by metal truss plates punched 

with barbs (that is, teeth) that are pressed into the truss members. Roof trusses are 

able to span the entire width of a home without interior support walls, allowing 

complete freedom in partitioning interior living space. The Metal Plate Connected 

Wood Truss Handbook contains span tables for typical truss designs (Callahan, 

2004).
Residential Structural Design Guide 5-44

FIGURE 5.8 Design Methods and Assumptions for a Sloped Roof Rafter

Roof truss manufacturers normally provide the required engineering 

design based on the loading conditions specified by the building designer or 

structural engineer of record (SER). The building designer is responsible for 

providing the following items to the truss manufacturer for design.

• Design loads.

• Truss profile.

• Support locations.

• Any special requirements.
Residential Structural Design Guide 5-45

The building designer must provide for permanent bracing of the truss 

system at locations designated by the truss designer to avoid invalidating any 

warranty on performance or design assumptions made by the truss designer. In 

general, such bracing may involve vertical cross-bracing, runners on the bottom 

chord, and the bracing of certain web members. In typical light-frame residential 

roof construction, properly attached roof sheathing provides adequate overall 

bracing of the roof truss system, and ceiling finishes typically provide lateral 

support to the bottom chord of the truss. The only exception is long web members 

that may experience buckling from excessive compressive loads. Section 5.6.6 

addresses gable-end wall bracing separately as it pertains to the role of the roof 

system in supporting the walls against lateral loads, particularly those produced 

by wind. For more information and details on permanent bracing of trusses, refer 

to Commentary for Permanent Bracing of Metal Plate Connected Wood Trusses

(WTCA, 1999). Temporary bracing during construction usually is the 

responsibility of the builder and is important for worker safety. Additional 

guidance on temporary bracing can be found in the Metal Plate Connected Wood 

Truss Handbook, pages 14-1 through 15-12 and appendix L (Callahan, 2004). 

Standard Practice for Metal Plate Connected Wood Truss Design Responsibilities

(WTCA, 2008) contains additional guidance on roles and responsibilities.

The National Design Standard for Metal Plate Connected Wood Truss 

Construction (ANSI/TPI 1-07) and commentary governs the design of trusses. 

Available from the Truss Plate Institute (TPI, 2007a and TPI, 2007b), ANSI/TPI 

1-07 includes the structural design procedure, requirements for truss installation 

and bracing, and standards for the manufacture of metal plate connectors. Truss 

plate manufacturers and truss fabricators generally have proprietary computerized 

design software based on ANSI/TPI 1-07, with modifications tailored to their 

particular truss-plate characteristics.

Cracking and separating of ceiling finishes may occur at joints between 

the walls and ceiling of roofs. In the unfavorable condition of high attic humidity, 

the top chord of a truss may expand while the lower roof members, typically 

buried under attic insulation, may not be similarly affected; thus, a truss may bow 

upward slightly. Other factors that commonly cause interior finish cracking but 

are not in any way associated with the roof truss include shrinkage of floorframing members, foundation settlement, or heavy loading of a long-span floor, 

resulting in excessive deflection that may “pull” a partition wall downward from 

its attachment at the ceiling. To reduce the potential for cracking of ceiling 

finishes at partition wall intersections, 2x wood blocking should be installed at the 

top of partition wall plates as a backer for the ceiling finish material (that is, 

gypsum board). Ceiling drywall should not be fastened to the blocking or to the 

truss bottom chord within 16 to 24 inches of the partition. Proprietary clips are 

available for use in place of wood blocking, and resilient metal “hat” channels 

may also be used to attach the ceiling finish to the roof framing.

Trusses also are frequently used for floor construction to obtain long spans 

and to allow for the placement of mechanical systems (for example, ductwork and 

sanitary drains) in the floor cavity. 
Residential Structural Design Guide 5-46

One efficient use of a roof truss is as a structural truss for the gable end 

above a garage opening to effectively eliminate the need for a garage door header. 

A truss often is used as a girder to support a large section of roof framing 

members including other roof trusses.

5.6.4 Roof Sheathing

Roof sheathing thickness typically is governed by the spacing of roofframing members and live or snow loads. Sheathing is typically installed in 

accordance with prescriptive sheathing span rating tables published in a building 

code or made available by manufacturers. If the limit of the prescriptive tables is 

exceeded, the designer may need to perform calculations; however, such 

calculations are rarely necessary in residential construction. The process of 

selecting rated roof sheathing is similar to that for floor sheathing.

Nails are the primary fasteners used to attach sheathing to roof rafters. The 

most popular nail types are sinker, box, and common, of which all have different 

characteristics that affect structural properties (refer to chapter 7). The IRC 

requires common nails but also permits the use of staples. Proprietary powerdriven fasteners (such as pneumatic nails and staples) are also used extensively. 

The building codes and APA tables recommend a fastener schedule for 

connecting sheathing to roof rafters. Generally, nails are placed at a minimum 6

inches on center at edges and 12 inches on center at intermediate supports. A 6-

inch fastener spacing should also be used at the gable-end framing to help brace 

the gable end. Nail size is typically 8d, particularly because thinner power-driven 

nails are most commonly used. Roof sheathing is commonly 7/16- to 5/8-inch 

thick on residential roofs. In some cases, shear loads in the roof diaphragm 

resulting from lateral loads (that is, wind and earthquake) may require a more 

stringent fastening schedule (refer to chapter 6 for a discussion of fastening 

schedules for lateral load design). More importantly, large suction pressures on 

roof sheathing in high wind areas (see chapter 3) require a larger fastener, closer 

spacing, or fasteners with greater withdrawal capacity, such as ring shank nails. In 

hurricane-prone regions, codes commonly require an 8d ring shank nail with 6-

inch, on-center spacing at all framing connections. At the gable-end truss or 

rafter, 4-inch fastener spacing is common.

While ring shank nails have greater withdrawal capacity, some types may 

also have reduced shear capacity and thus may require closer spacing to resist 

shear in the diaphragm. This subject is covered further in the International Code 

Council’s Standard on Residential Construction in High Wind Regions, ICC 600 

(ICC, 2008).

5.6.5 Roof Overhangs

Overhangs are projections of the roof system beyond the exterior wall line 

at either the eave or the rake (the sloped gable end). Overhangs protect walls from 

rain and shade windows from direct sun. When a roof is framed with wood 

trusses, an eave overhang typically is constructed by extending the top chord 
Residential Structural Design Guide 5-47

beyond the exterior wall. When a roof is framed with rafters, the eave overhang is 

constructed by using rafters that extend beyond the exterior wall. The sloped 

rafters are cut with a “bird-mouth” to conform to the bearing support. Gable-end 

overhangs usually are framed by using a ladder panel that cantilevers over the 

gable end for either stick-framed or truss roofs (refer to figure 5.9 for illustrations 

of various overhang constructions).

A study completed in 1978 by the Southern Forest Experiment Station for 

the U.S. Department of Housing and Urban Development found that the 

protection afforded by overhangs extends the life of the wall below, particularly if 

the wall is constructed of wood materials (HUD, 1980). Entitled the Prevention 

and Control of Decay in Homes, the report correlates the climate index of a 

geographic area with a suggested overhang width and recommends highly 

conservative widths. Moisture Resistant Homes (HUD, 2006) also is a good 

source of information on the issue of moisture and decay in residential structures. 

As a reasonable guideline (given that in many cases no overhang is provided), 

protective overhang widths should be 12 to 24 inches in humid climates—more, if 

practicable. A rule of thumb is to provide a minimum of 12 inches of overhang 

width for each story of protected wall below; however, overhang width can 

significantly increase wind uplift loads on a roof, particularly in high-wind 

regions. The detailing of overhang framing connections (particularly at the rake 

overhang on a gable end) is a critical consideration in hurricane-prone regions. 

Often, standard metal clips or straps provide adequate connection. The need for 

special rake overhang design detailing depends on the length of the overhang, the 

design wind load condition, and the framing technique that supports the overhang 

(that is, 2x outriggers versus cantilevered roof sheathing supporting ladder 

overhang framing).
Residential Structural Design Guide 5-48

In high-wind areas, soffits should be solidly nailed to framing to prevent 

wind uplift of the soffit covering or to prevent the soffit covering from being 

pulled away from the framing. This will likely require solid blocking along the 

exterior wall and between roof rafters, or trusses that frame the overhang to 

provide nailing ledgers for the soffit covering. The 2012 IRC and ICC-600 also 

contain prescriptive details for constructing a blocking panel consisting of wall 

sheathing and 2x4 framing that can be substituted for solid blocking. Truss 

manufacturers also can provide a truss component in lieu of solid blocking or a 

prescriptive blocking panel.

5.6.6 Gable-End Wall Bracing

Roof framing provides lateral support to the top of the walls where trusses 

and rafters are attached to the wall top plate. Likewise, floor framing provides 

lateral support to the top and bottom of walls, including the top of foundation 

walls. At a gable end, when platform framing is used (see section 1.1), however, 

the top of the wall is not directly connected to roof-framing members; instead, it 

FIGURE 5.9 Typical Roof Overhang Construction
Residential Structural Design Guide 5-49

is attached to the bottom of a gable-end truss, and lateral support at the top of the 

wall is provided by the ceiling diaphragm. In higher wind regions, the joint may 

become a “hinge” if it is not braced. Accordingly, common practice is to brace the 

top of the end wall (or bottom of the gable-end roof framing) with 2x4 or 2x6 

framing members that slope upward to the roof diaphragm to attach to a blocking 

or a ridge “beam,” as shown in figure 5.10. Alternatively, braces may be laid flat 

on ceiling joists or truss bottom chords to transfer the load to a ceiling diaphragm. 

Given that braces must transfer inward and outward forces resulting from positive 

wind pressure or suction on the gable-end wall, the braces commonly are attached 

to the top of the gable-end wall with straps to transfer tension forces that may 

develop in hurricanes and other extreme wind conditions. The need for special 

detailing of gable-end wall braces depends on the height and area of the gable end 

(that is, tributary area) and the design wind load. The gable-end wall also can be 

braced using a wood structural panel attached to the gable-end framing and the 

ceiling framing members. Specific details for bracing gable-end walls in highwind regions can be found in the Wood Frame Construction Manual (AWC, 

2012c) or the ICC 600 Standard for Residential Construction in High Wind 

Regions (ICC, 2008).

As an alternative to the preceding strategy, the gable-end wall may be 

framed with continuous studs that extend to the roof sheathing at the gable end 

(that is, balloon framed). If the gable-end wall encloses a two-story room—such 

as a room with a cathedral ceiling, the studs must extend to the roof sheathing; 

otherwise, a hinge is created in the wall that could cause cracking of wall finishes 

(even in a moderate wind) and could easily precipitate failure of the wall in an 

extreme wind. Depending on wall height, stud size, stud spacing, and the design 

wind load condition, taller, full-height studs may have to be enlarged to meet 

deflection or bending capacity requirements. The designer must exercise 

judgment in this framing application with respect to deflection criteria.

Finally, as an alternative that avoids the gable-end wall-bracing problem, a 

hip roof may be used. The hip shape is inherently more resistant to wind damage 

in hurricane-prone wind environments and braces the end walls against lateral 

wind loads by directly attaching to rafters.
Residential Structural Design Guide 5-50

5.7 Design Examples

In this section, a number of design examples illustrate the design of 

various elements discussed in this chapter. The examples are intended to also 

provide practical advice; therefore, they are embellished with numerous notes and 

recommendations to improve the practicality and function of various possible 

design solutions. The examples also are intended to promote the designer’s 

creativity in arriving at the best possible solution for a particular application.

FIGURE 5.10 Gable-End Wall Bracing

GABLE-END WALL 

BRACE (AS REQ’D)

GABLE-END WALL

LADDER FRAMING

HORIZONTAL 

BRACE/RUNNER 

(AS REQ’D)
Residential Structural Design Guide 5-51

EXAMPLE 5.1 Typical Simple Span Floor Joist Design

Given Live load (L) = 30 psf (bedroom area)

Dead load (D) = 10 psf

Trial joist spacing = 16 on center

Trial joist size = 2x8

Trial joist species and grade = Hem-Fir, No. 1 (S-dry, 19% MC)

Find Maximum span for specified joist member.

Solution

1. Determine tabulated design values by using NDS-S (tables 4A and 1B).

Fb = 975 psi Ixx = 47.63 in4

Fv = 75 psi Sxx = 13.14 in3

Fc⊥ = 405 psi b = 1.5 in

E = 1,500,000 psi d = 7.25 in

2. Lumber property adjustments and adjusted design values (section 5.2.4 and 

NDS•2.3)

CD = 1.0 (section 5.2.4.1)

Cr = 1.15 (table 5.4)

CF = 1.2 (NDS-S table 4A adjustment factors)

CH = 2.0 (section 5.2.4.3)

CL = 1.0 (NDS•3.3.3, continuous lateral support)

Cb = 1.0 (NDS•2.3.10)

Fb' = FbCrCFCDCL = 975 (1.15)(1.2)(1.0)(1.0) = 1,345 psi

Fv' = FvCHCD = 75 (2)(1.0) = 150 psi

Fc⊥' = Fc⊥Cb = 405 (1.0) = 405 psi

E' = E = 1,500,000 psi

3. Calculate the applied load.

W = (joist spacing)(D+L) = (16 in)(1 ft/12 in)(40 psf) = 53.3 plf

4. Determine maximum clear span based on bending capacity

Mmax = w

2

8 = (53.3 plf)(

2)

8 = 6.66 2

fb = = = 6.08 2

fb ≤ Fb'

6.08 2 ≤ 1,345 psi

2 = 221

 = 14.9 ft = 14 ft-11 in (maximum clear span resulting from bending 

stress)

S

M

3

2

13.14 in

) ft )(12 in (6.66
Residential Structural Design Guide 5-52

5. Determine maximum clear span based on horizontal shear capacity

Vmax = = = 26.7 

fv = = = 3.7

fv ≤ Fv'

3.7 ≤ 150 psi

 = 40.5 ft = 40 ft-6 in (maximum clear span resulting from

horizontal shear stress)

6. Determine maximum clear span based on bearing capacity

Bearing length = (3.5-in top plate width) - (1.5-in rim joist width) = 2 in

fc⊥ = = = 8.9

fc⊥ < Fc⊥'

8.9 ≤ 405 psi

 = 45.5 ft = 45 ft-6 in (maximum clear span resulting from 

bearing stress)

7. Consider maximum clear span based on deflection criteria (section 5.3.2)

max = = = 1.26 x 10-5

4

*

applied live load of 30 psf only

all = (12 in/ft) = 0.033

max ≤ all

1.26 x 10-5 4 ≤ 0.033

3 = 2,619

= 13.8 ft = 13 ft-10 in (recommended clear span limit

resulting from deflection criteria)

8. Consider floor vibration (section 5.3.2).

The serviceability deflection check was based on the design floor live load for 

bedroom areas of 30 psf. The vibration control recommended in section 5.3.2 is to 

use a 40 psf design floor live load with the /360 deflection limit. Given that the 

span will not be greater than 15 feet, using the absolute deflection limit of 0.5 inch 

is not necessary.

w = (16 in)(1 ft/12 in)(40 psf) = 53.3 plf

all = = 0.033

2

w

2

(53.3 plf)() 

2A

3V

















(1.5 in)(7.25 in)

26.7

2

3 



Ab

w

2

1 

(2 in)(1.5 in)

(53.3 plf)( ) 2

1 



384EI

5w

4  ( )

384 (1,500,000 psi)(47.63 in )

5 (40 plf) (1,728 in / ft )

4

* 4 3 3 

360





 





)(12 in / ft) 360 (  
Residential Structural Design Guide 5-53

max = = = 1.7 x 10-5 4

*

applied live load of 40 psf only

max ≤ all

1.7 x 10-5 4 ≤ 0.033

3 = 1,941

 = 12.5 ft = 12 ft-6 in (recommended clear span limit resulting from 

vibration)

Conclusion The serviceability limit states used for deflection and floor vibration limit the 

maximum span. The deflection-limited span is 13 ft-10 in, and the vibrationlimited span is 12 ft-6 in. Span selection based on deflection or vibration is an issue 

of designer judgment. The maximum span limited by the structural safety checks 

was 14 ft-11 in resulting from bending; therefore, the serviceability limit will 

provide a notable safety margin above that required. No. 2 grade lumber should be 

considered for economy because it will have only a small effect on the 

serviceability limits. Conversely, if floor stiffness is not an expected issue with the 

owner or occupant, the span may be increased beyond the serviceability limits if 

needed to “make it work.” Many serviceable homes have been built with 2x8 floor 

joists spanning as much as 15 feet; however, if occupants have a low tolerance for 

floor vibration, a lesser span should be considered.

For instructional reasons, shrinkage across the depth of the floor joist or floor 

system may be estimated as follows, based on the equations in section 5.3.2:

d1 = 7.25 in M1 = 19% maximum (S-dry lumber)

d2 = ? M2 = 10% (estimated equilibrium MC)

d2 = d1 = 7.25 in = 7.1 in

Shrinkage ≅ 7.25 ft-7.08 in = 0.15 in (almost 3/16 in)

In a typical wood-framed house, shrinkage should not be a problem, provided that 

the shrinkage is uniform throughout the floor system. In multistory platform frame 

construction, the same amount of shrinkage across each floor can add up to become 

a problem, and mechanical systems and structural details should allow for such 

movement. Kiln-dried lumber may be specified to limit shrinkage and building 

movement after construction.

384 EI

5 w

4 

384(1.5 x10 psi)(47.63 in )

5(53.3 plf) ( ) (1,728 in / ft )

6 4

* 4 3 3 



 

























− −

− −

100

a 0.2M

1

100

a 0.2M

1

1

2

























− −

− −

100

6.031 0.2 (19) 1

100

6.031 0.2 (10) 1
Residential Structural Design Guide 5-54

EXAMPLE 5.2 Simple Span Floor Joist Design (Optimum Lumber)

Given Live load (L) = 40 psf 

Dead load (D) = 10 psf

Clear span = 14 ft-2 in

Joist size = 2x10

Find Optimum lumber species and grade

Solution

1. Calculate the applied load.

W = (joist spacing)(D+L) = (2 ft)(40 psf +10 psf) = 100 plf

2. Determine bending stress

Mmax = w

2

8 = (100 plf)(14.17 ft2)8 = 2,510 ft-lb

Fb = = = 1,408 psi

3. Determine horizontal shear stress.

Vmax = = = 709 lb

fv = = = 77 psi

4. Determine bearing stress.

R1 = R2 = Vmax = 709 lb

fc⊥ = = = 236 psi

=

Wall and roof loads, if any, are carried through rim/band joist.

5. Determine minimum modulus of elasticity resulting from selected deflection 

criteria.

max = =

= 733,540/E 

*includes live load of 40 psf only

all ≤

max ≤ all

= 

Emin = 1.55 x 106 psi 

S

M

3 21.39 in

(2,510 ft − lb)(12 in / ft)

2A

3V

2 (1.5 in)(9.25 in)

3(709 lb)

Ab

R

(2 in)(1.5 in)

709 lb

384E (98.93 in )

5 (80 plf)*(14.17 ft) (1,728in / ft )

4

4 3 3

360



E

733,540

360

(14.17 ft)(12 in / ft)

2

w

2

(100 plf)(14.17 ft)

5wl

4

 

384EI
Residential Structural Design Guide 5-55

6. Determine minimum modulus of elasticity resulting from vibration.

The span required is not greater than 15 feet, and the /360 deflection check uses a 

40 psf floor live load; therefore, the deflection check is assumed to provide 

adequate vibration control.

7. Determine minimum required unadjusted properties by using NDS tabulated 

lumber data.

Bending fb ≤ Fb'

Fb' = FbCrCFCD

Fbmin = = 1,113 psi

Horizontal shear fv ≤ Fv'

Fv' = FvCHCD

Fvmin = = = 39 psi

Bearing fc⊥ ≤ Fc⊥' (assume minimum 2-in bearing)

Fc⊥ = Fc⊥Cb

Fc⊥min = = 236 psi

Minimum unadjusted tabulated properties required

Fb = 1,113 psi Fc⊥ = 236 psi

Fv = 39 psi E = 1.55x106 psi

8. Select optimum lumber grade considering local availability and price by using 

NDS-S table 4A or 4B data.

Minimum No. 2 grade lumber is recommended for floor joists because of factors 

related to lumber quality, such as potential warping and straightness, that may 

affect constructability and create call-backs.

Considering 2x10 Douglas-Fir-Larch, the grade below (No. 1 and better) was 

selected to meet the required properties.

Fb = 1,200 psi > 1,113 psi OK

Fv = 95 psi > 39 psi OK

Fc⊥ = 625 psi > 236 psi OK

E = 1.8x106 psi > 1.55x106 psi OK

Conclusion Many other species and grades should be considered, depending on local 

availability and cost. Also, the No. 1 and higher grades are generally considered 

“premium” lumber. A lower grade 2x12 should be considered or, perhaps, 

engineered wood I-joists.

(1.15)(1.1)(1.0)

1,408 psi

H D

v

C C

f

(2)(1.0)

77 psi

(1.0)

f c⊥
Residential Structural Design Guide 5-56

EXAMPLE 5.3 Built-Up Floor Girder Design

Given Loads

Floor live load (L) = 40 psf

Floor dead load (D) = 10 psf

Required girder span (support column spacing) = 14 ft

Joist span (both sides of girder) = 12 ft

Species = Southern Pine, No. 1

Maximum girder depth = 12

Find Minimum number of 2x10 or 2x12 required for the built-up girder.

Solution

1. Calculate the design load.

W = (Trib. floor joist span)(D + L) = (12 ft)(40 psf + 10 psf) = 600 plf

2. Determine tabulated design values (Revised values effective 6/1/13)

Fb = 1000 psi FC⊥ = 565 psi

Fv = 175 psi E = 1.6x106 psi

3. Lumber property adjustments (section 5.2.4):

Cr = 1.2 (table 5.4) CD = 1.0

CF = 1.0 Cb = 1.0

CH = 2.0 CL = 1.0

(Compression flange laterally braced by connection of floor joists to top or side of 

girder)

Fb' = FbCDCrCFCL = 1,000 psi (1.0)(1.2)(1)(1) = 1,200 psi

FV' = FlCDCH = 175 psi (1.25)(2.0) = 438 psi

Fc⊥' = Fc⊥Cb = 565 psi (1) = 565 psi

E' = E = 1.6x106 psi

4. Determine number of members required resulting from bending.

Mmax = = = 14,700 ft-lb

fb = = =

fb ≤ Fb'

≤ 1,200 psi

Sx = 147 in3

8

w

2 

8

(600plf)(14 ft)

2

S

M

S

(14,700 ft − lb)(12 in / ft)

S

176,400

S

176,400
Residential Structural Design Guide 5-57

Using table 1B in NDS-S

five 2x12 S = 5 (31.64) = 158 > 147 OK, but five pieces is too many

5. Determine number of members required resulting from horizontal shear

Vmax = = = 4,200 lb

fv = = =

fv ≤ Fv'

≤ 438 psi

A = 14 in2 1 2x12 A = 16.9 > 14 OK

1 2x10 A = 13.9 ≈ 14 OK

6. Determine required bearing length using 4 2x12.

R1 = R2 = Vmax = 4,200 lb

fc⊥ = = =

fc1 ≤ Fc⊥'

≤ 565 psi

b = 1.24 in OK

7. Determine member size resulting from deflection.

max = = = 

*includes 40 psf live load only

all ≤ = = 0.47 in

max ≤ all

= 0.47 in

EI = 8.8 x 108 = (1.6 x 106

)(I)

I = 550 in4

 select four 2x12 ( I = 178 in4)

I = 712 in4 > 550 OK

8. Check girder for floor system vibration control (see section 5.3.2).

Girder span, 1 = 14 ft

Joist span, 2 = 12 ft

TOTAL = 26 ft > 20 ft

2

w

2

600 plf (14 ft)

2A

3V 



 





A

4200

2

3

A

6,300 lb

A

6,300 lb

Ab

R

(6 in)( )

4,200 lb

b  b

700



b

700



384EI

5w

4 

384 EI

5 (480 plf)*(14 ft) (1,728 in / ft ) 4 3 3

EI

4.15 x108

360



360

14 ft (12 in / ft)

EI

4.15 x108
Residential Structural Design Guide 5-58

Therefore, check girder using or to stiffen floor system.

Try 

max = (as before)

all = = = 0.35 in

max ≤ all

= 0.35 in

EI = 1.2 x 109

I = = 750 in4

Using table 1B in NDS, use

five 2x12 I = 5 (178 in4) = 890 in4 > 750 in4 OK

Conclusion The bending stress limits the floor girder design to five 2x12 (No. 1, Southern 

Yellow Pine). The use of five 2x12 also provides a “stiff” girder with respect to 

floor vibration (i.e., deflection limit of . As a practical alternative, a steel 

“floor beam” (for example, W-shape) or an engineered wood beam may also be 

used, particularly if “clearance” is a concern.

480



600



600



480



EI

4.15 x108

480



480

14 ft (12 in / ft)

EI

4.15 x108

6

9

1.6 10

1.2 10

x

x

480


Residential Structural Design Guide 5-59

EXAMPLE 5.4 Exterior Bearing Wall Design

Given Stud size and spacing = 2x4 at 24 in on center

Wall height = 8 ft

Species and grade = Spruce-Pine-Fir, Stud Grade

Exterior surface = 7/16-in-thick OSB

Interior surface = 1/2-in-thick, gypsum wall board

Wind load (100 mph, gust) = 16 psf (see chapter 3, example 3.2)

Find Vertical load capacity of stud wall system for bending (wind) and axial 

compression (dead load) and for axial compression only (that is, dead, live, and 

snow loads); refer to chapter 3, table 3.1, for applicable load combinations.

Wall Loading Diagram

Solution

1. Determine tabulated design values for the stud by using the NDS-S (table A4).

Fb = 675 psi Fc⊥ = 425 psi

Ft = 350 psi Fc = 725 psi

Fv = 70 psi E = 1.2x106 psi

2. Determine lumber property adjustments (see section 5.2.4).

CD = 1.6 (wind load combination)

= 1.25 (gravity/snow load combination)

Cr = 1.5 (sheathed wall assembly, table 5.4)

CL = 1.0 (continuous lateral bracing)

CF = 1.05 for Fc

= 1.1 for Ft

= 1.1 for Fb

3. Calculate adjusted tensile capacity.

Not applicable to this design. Tension capacity is OK by inspection.

4. Calculate adjusted bending capacity.

Fb' = FbCDCLCFCr = (675)(1.6)(1.0)(1.1)(1.5) = 1,782 psi
Residential Structural Design Guide 5-60

5. Calculate adjusted compressive capacity (NDS•3.7).

Fc

* = FcCDCF = (725 psi)(1.6)(1.05) = 1,218 psi

E' = E = 1.2x106 psi

KcE = 0.3 visually graded lumber

c = 0.8 sawn lumber

FcE = = = 479 psi

Cp = – (column stability factor)

= – = 0.35

Fc' = FcCDCFCP = (725 psi)(1.6)(1.05)(0.35) = 426 psi

Axial load only case

Calculations are same as above except use CD = 1.25

Fc* = 952 psi

Cp = 0.44

Fc' = FcCDCFCP = 725 psi (1.25)(1.05)(0.44) = 419 psi

6. Calculate combined bending and axial compression capacity for wind and gravity 

load (dead load only) by using the combined stress interaction (CSI) equation 

(NDS•3.9.2).

fb = =

=

= 1,004 psi

+ ≤ 1.0 (CSI equation for bending in strong axis of stud 

only)

2

e

cE

d

l

K E'





 



 2

6

3.5 in

8 ft (12 in / ft)

0.3(1.2 x10 psi)



 

2c

* F

F 1 c

cE 



 



 +

c

F *

F

2c

F

F 1

c

cE

2

*

c

cE

−













































+

2 (0.8)

1,218

479 1 



 



 +

0.8

1,218

479

2 (0.8)

1,218

479 1

2

−



















 



 



 +

S

M

S

w8

1 2 

[ ]

3

2

3.06 in

) (1 ft /12 in) ft 8 ft)(12 in (24 in)(16 psf) ( 8

1

2

c

c

F '

f





















 



 − cE1

c b

b

F

f F ' 1

f
Residential Structural Design Guide 5-61

+ = 1.0 (solve CSI equation for fc)

fc, max = 163 psi/stud

P = fcA = (163 psi/stud)(1.5 in)(3.5 in) = 856 lb/stud

w = (856 lb/stud) = 428 plf (uniform dead load at top of wall)

Therefore, the maximum axial (dead) load capacity is 428 plf with the wind load 

case (that is, D+W).

7. Determine maximum axial gravity load without bending load.

This analysis applies to the D + L + 0.3(S or Lr) and D + (S or Lr) + 0.3L load 

combinations (see table 3.1, chapter 3).

Using Fc' determined in step 5 (axial load only case), determine the stud capacity 

acting as a column with continuous lateral support in the weak-axis buckling 

direction.

Fc ≤ Fc'

≤ 419 psi

Pmax = (419 psi)(1.5 in)(3.5 in) = 2,200 lbs/stud

Maximum axial load capacity (without simultaneous bending load) is 2,200 

lbs/stud or 1,100 lbs/lf of wall.

8. Check bearing capacity of wall plate.

Not a capacity limit state (Fc⊥ is based on deformation limit state, not actual 

bearing capacity). OK by inspection.

Conclusion The axial and bending load capacity of the example wall is ample for most 

residential design conditions; thus, in most cases, using the prescriptive stud tables 

found in residential building codes may save time. Only in very tall walls (that is, 

greater than 10 feet) or more heavily loaded walls than typical will a special 

analysis as shown here be necessary, even in higher wind conditions. The 

controlling factor will likely be a serviceability limit state (that is, wall deflection) 

rather than strength, as shown in several of the floor design examples. In such 

cases, the wall system deflection adjustment factors of table 5.6 should be 

considered.

Note:

The axial compression capacity determined in this example is conservative because 

the actual EI (the product of the modulus of elasticity and the moment of inertia) of 

the wall system is not considered in the determination of Cp for stability. No 

method is currently available to include system effects in the analysis of Cp; 

however, a Ke factor of 0.8 may be used as a reasonable assumption to determine 

the effective buckling length, e, which is then used to determine Cp (see 

NDS•3.7.1).

2

c

426

f







 











 



 − 479

f 1,782 1

1,004

c

















2 ft

1 stud

A

P
Residential Structural Design Guide 5-62

EXAMPLE 5.5 Header System Design

Given Two-story house

Required header span = 6.3 ft (rough opening)

Species and grade = Spruce-Pine-Fir (south), No. 2 Grade

Loads on first-story header

wfloor = 600 plf (includes floor dead and live loads)

wwall = 360 plf (includes dead, live, and snow loads supported by wall 

above header)*

wtotal = 960 plf (includes dead, live, and snow loads)*

*Combined loads are determined in accordance with table 3.1 of chapter 3.

Find Determine header size (2x8 or 2x10) by considering system effect of all horizontal 

members spanning the opening.

Header System

Solution

1. Determine tabulated design values by using the NDS-S (table 4A).

Fb = 775 psi

Fv = 70 psi

Fc⊥ = 335 psi

E = 1.1x106 psi

2. Determine lumber property adjustments (section 5.2.4).

Cr = 1.3 (2x10 double header per table 5.8)

= 1.2 (2x8 double header per table 5.4)

CD = 1.25 (snow load)

CF = 1.1 (2x10)

= 1.2 (2x8)

CH = 2.0

Cb = 1.0

CL = 1.0 laterally supported

Fb' = FbCDCrCFCL = (775 psi)(1.25)(1.3)(1.1)(1.0) = 1,385 psi [2x10]

= (775 psi)(1.25)(1.2)(1.1)(1.0) = 1,279 psi [2x8]

Fv' = FvCDCH = (70 psi)(1.25)(2) = 175 psi

Fc⊥' = F c⊥Cb = (335psi)(1) = 335 psi

E' = E = 1.1x106 psi

With double top plate, Fb can be increased by 5 percent (table 5.8)
Residential Structural Design Guide 5-63

Fb' = Fb' (1.05) = 1,385 psi (1.05) = 1,454 psi [2x10]

Fb' = Fb' (1.05) = 1,279 psi (1.05) = 1,343 psi [2x8]

3. Determine header size resulting from bending for floor load only.

Mmax = = = 3,169 ft-lb

fb = ≤ Fb'

1,454 psi = 

S = 26.2 in3

S for two 2x10 = 2(21.39 in3

) = 42.78 in3 >

26.2 in3 OK

Try two 2x8

1,343 psi =

S = 28.3 in3

S for two 2x8 = 2 (13.14) = 26.3 in3 < 28.3 in3

(close, but no good)

4. Determine member size resulting from bending for combined floor and supported 

wall loads by using the 1.8 system factor from table 5.8 but not explicitly 

calculating the load sharing with the band joist above.

Fb' = Fb (CD)(Cr)(CF)(CL) = 775 psi 

(1.25)(1.8)(1.1)(1.0) = 1,918 psi

Mmax = = =

5,070 ft-lb

fb = ≤ Fb'

1,918 psi=

S = 31.7 in3

S for two 2x10 = 42.78 in3 > 31.7 in3 OK

5. Check horizontal shear.

Vmax = = = 1,950 lb

fv = = = 106 psi

fv ≤ Fv'

106 psi < 175 psi OK

6. Check for adequate bearing.

8

w

2 

8

(600 plf)(6.5 ft)

2

S

Mmax

S

3,169 ft − lb (12 in / ft)

S

3,169 ft − lb (12 in / ft)

8

w

2 

8

(360 plf 600 plf)(6.5 ft)

2 +

S

M

S

5,070 ft − lb (12 in / ft)

2

w

2

(600 plf)(6.5)

2A

3V

2 (2)(1.5 in)(9.25 in)

3(1,950 lb)
Residential Structural Design Guide 5-64

R1= R2 = Vmax = 1,950 lb

fc⊥= = =

fc⊥≤ Fc⊥'

= 335

b = 1.9 in

OK for bearing, use two 2x4 jack studs (b = 3 in)

7. Check deflection.

max = = = 0.11 in

all = L/240 = 0.325 in

max < all

Conclusion Using a system-based header design approach, a two-2x10 header of No. 2 SprucePine-Fir is adequate for the 6 ft-3 in span opening. The loading condition is 

common to the first story of a typical two-story residential building. Using a 

stronger species or grade of lumber would allow the use of a two-2x8 header. 

Depending on the application and potential savings, using the header tables found 

in a typical residential building code may be more cost effective. For cost-effective 

ideas and concepts that allow for reduced header loads and sizes, refer to Cost 

Effective Home Building: A Design and Construction Handbook (NAHBRC, 

1994). The document also contains convenient header span tables. For headers that 

are not part of a floor–band joist system, the design approach of this example is 

still relevant and similar to that used for floor girders. The 1.8 system factor used 

here would not apply, however, and the double top plate factor would apply only as 

appropriate.

Ab

R

(2)(1.5 in)( )

1,950 lb

b  b

650



b

650



384EI

5w

4 

384 (1.1x10 psi)[(98.9 in )(2)])

5 (600 plf)(6.5 ft) (1,728 in / ft )

6 4

4 3 3

240

(6.5 ft)(12 in / ft)
Residential Structural Design Guide 5-65

EXAMPLE 5.6 Column Design

Given Basement column supporting a floor girder

Spruce-Pine-Fir, No. 2 Grade

Axial design load is 4,800 lbs (D + L)

Column height is 7.3 ft (unsupported)

Find Adequacy of a 4x4 solid column

Solution

1. Determine tabulated design values by using the NDS-S (table 4A).

Fc = 1,150 psi

E = 1.4x106 psi

2. Lumber property adjustments (section 5.2.4):

CD = 1.0

CF = 1.15 for Fc

3. Calculate adjusted compressive capacity (NDS•3.7).

Trial 4x4

Fc* = FcCDCF = 1,150 psi (1.0)(1.15) = 1,323 psi

E' = E = 1.4x106 psi

KcE = 0.3 for visually graded

c = 0.8 for sawn lumber

FcE = = = 670 psi

Cp = –

= – = 0.44

Fc' = FcCDCFCp = (1,150 psi)(1.0)(1.15)(0.44) = 582 psi

Pall = F'cA = (582 psi)(3.5 in)(3.5 in) = 7,129 lb > 4,800 lb

OK

Conclusion A 4x4 column is adequate for the 4,800-pound axial design load and the stated 

height and support conditions; in fact, greater column spacing could be used. Note 

that the analysis was performed with a solid sawn column of rectangular 

dimension. If a nonrectangular column is used, buckling must be analyzed in the 

weak-axis direction in consideration of the distance between lateral supports, if 

2

e

cE

d

K E'





 



 2

6

3.5 in

7.3 ft (12 in / ft)

0.3(1.4 x10 psi)





 





2c

* F

F 1 c

cE 



 



 +

c

* Fc

F

2c

* F

F 1 cE

2

c

cE

−























 



 



 +

2 (0.8)

1,323

670 1 



 



 +

0.8

1,323

670

2 (0.8)

1,323

670 1

2

−



















 



 



 +
Residential Structural Design Guide 5-66

any, in that direction. If a built-up column is used, it is NOT treated the same way 

as a solid column. Even if the dimensions are nearly the same, the built-up column 

is more susceptible to buckling resulting from slippage between adjacent members 

as flexure occurs in response to buckling (only if unbraced in the weak-axis 

direction of the built-up members). Slippage depends on how well the built-up 

members are fastened together, which is accounted for by the use of an additional 

adjustment (reduction) factor applied to the Cp equation (see section 5.5.5 and 

NDS•15.3).
Residential Structural Design Guide 5-67

EXAMPLE 5.7 Simply Supported, Sloped Rafter Design

Given Two-story home

Rafter spacing 16 in on center

Rafter horizontal span is 12 ft (actual sloped span is 14.4 ft)

8:12 roof slope

Design loads (see chapter 3):

Dead load = 10 psf

Roof snow load = 20 psf (20 psf ground snow)

Wind load (90 mph, gust) = 12.7 psf (outward, uplift)

= 7.4 psf (inward)

Roof live load = 10 psf

Find Minimum rafter size using No. 2 Douglas-Fir-Larch (refer to figure 5.7 for load 

diagram).

Solution

1. Evaluate load combinations applicable to rafter design (see chapter 3, table 3.1).

The load combinations to consider and initial assessment based on the magnitude 

of the given design loads follows—

D + (Lr or S) Controls rafter design in inward-bending direction 

(compression side of rafter laterally supported); Lr can be 

ignored because the snow load magnitude is greater.

0.6D + Wu May control rafter design in outward-bending direction 

because the compression side now has no lateral bracing 

unless specified; also important to rafter connections at the 

bearing wall and ridge beam.

D + W Not controlling by inspection; gravity load D + S controls in 

the inward-bending direction.

2. Determine relevant lumber property values (NDS-S, table 4A).

Fb = 900 psi

Fv = 95 psi

E = 1.6 x 106 psi

3. Determine relevant adjustments to property values, assuming a 2x8 will be used 

(section 5.2.4).

CD = 1.6 (wind load combinations)

= 1.25 (snow load combination)

Cr = 1.15 (2x8, 24 inches on center)

CH = 2.0

CF = 1.2 (2x8)

CL = 1.0 (inward bending, D + S, laterally braced on compression edge)

= 0.32 (outward bending, 0.6 D + W, laterally unbraced on compression 

edge)*

*

Determined in accordance with NDS•3.3.3
Residential Structural Design Guide 5-68

e = 1.63 u + 3d

= 1.63 (14.4 ft) + 3 (7.25 in)(1 in/12ft)

= 25.3 ft

RB = =

= 31 < 50 (OK)

KbE = 0.439 (visually graded lumber)

FbE = = = 730 psi

Fb* = FbCDCrCF

= 900 psi (1.6)(1.15)(1.2) = 1,987 psi

CL = -

CL = 0.36 (2x8)

4. Determine rafter transverse bending load, shear, and moment for the wind uplift 

load case (using method A of figure 5.8).

The wind load acts transverse (that is, perpendicular) to the rafter; however, the 

snow load acts in the direction of gravity and must be resolved to its transverse 

component. Generally, the axial component of the gravity load along the rafter 

(which varies, depending on end connectivity) is ignored and has negligible impact 

considering the roof system effects that also are ignored. The limited overhang 

length will have a negligible impact on the design of the rafter; thus, the rafter can 

be reasonably analyzed as a sloped, simply supported bending member. In 

analyzing wind uplift connection forces at the outside bearing of the rafter, the 

designer should consider the additional uplift created by the small overhang, 

although for the stated condition it would amount only to about 20 pounds 

additional uplift load.

The net uniform uplift load perpendicular to the rafter is determined as follows:

wD, transverse = wD (cos θ)

= (10 psf)(1.33 ft)(cos 33.7°)

= 11 plf

ww, transverse = (12.7 psf)(1.33 ft) = 17 plf (uplift)

wtotal, transverse = 17 plf-11 plf = 6 plf (net uplift)

Shear, Vmax = = = 44 lbs

Moment, Mmax = 1/8 w2 

= 1/8 (6 plf)(14.4 ft)2 = 156 ft-lb

5. Determine bending load, shear, and moment for the gravity load case (D + S) using 

method B of figure 5.8 (horizontal span).

wD = (10 psf)(14.4 ft)(1.33 ft)/12 ft-horizontal = 16 plf

wS = (20 psf)(12 ft)(1.33 ft)/12 ft-horizontal = 27 plf

wtotal = 43 plf

wtotal = (43 plf)(cos 33.7°) = 36 plf

2

e

b

 d

2 (1.5 in)

(25.5 ft)(12 in / ft)(7.25 in)

2

B

bE

R

K E'

2

6

(31)

0.439 (1.6x10 psi)

1.9

1 (F / F *) + bE b

0.95

F / F *

1.9

1 (F / F *) bE b

2

bE b − 



 



 +

2

w

2

(6 plf)(14.4 ft)
Residential Structural Design Guide 5-69

Shear, Vmax = = 216 lb

Moment, Mmax = 1/8 (36 plf)(12 ft)2 = 648 ft-lb

6. Check bending stress for both loading cases and bending conditions.

Outward Bending (0.6D + Wu)

fb =

= (12 in/ft) = 142 psi

Fb' = FbCDCrCFCL

= 900 psi (1.6)(1.15)(1.2)(0.36) = 715 psi

fb << Fb' OK, 2x8 works, and lateral bracing of bottom compression 

edge is not required.

Inward Bending (D + S)

fb =

= (12 in/ft) = 591 psi

Fb' = FbCDCrCFCL

= 900 psi (1.25)(1.15)(1.2)(1.0) = 1,553 psi

fb << Fb' OK

7. Check horizontal shear.

Vmax = 216 lb (see step 5)

fv = = = 30 psi

Fv' = FvCDCH = 95 psi (1.25)(2.0) = 238 psi

fv << Fv' OK

8. Check bearing.

OK by inspection.

9. Check deflection criteria for gravity load condition (section 5.2.2).

all = = = 1.0 in

max = = (1,728 in3

/ft3)

= 0.4 in

max << all OK, usually not a mandatory roof check

Conclusion A 2x8, No. 2 Douglas-Fir-Larch rafter spaced at 16 inches on center was shown to 

have ample capacity and stiffness for the given design conditions. In fact, using 

2

(36 plf)(12 ft)

2

(36 plf)(12 ft)

S

M

3 13.14 in

156 ft − lb

S

M

3 13.14 in

648 ft − lb

2A

3V

2 (1.5 in)(7.25 in)

3(216 lb)

180



180

(14.4 ft)(12 in / ft)

384 EI

5w

4 

384 (1.6 x10 psi)(47.6 in )

5 (36 plf)(14.4 ft)

6 4

4
Residential Structural Design Guide 5-70

19.2-inch on center spacing (that is, five joists every 8 feet) would also work and 

be a more efficient use of lumber. A 2x6 could also possibly result in a reasonable 

rafter design for this application. For other concepts in value-added framing design, 

consult Cost Effective Home Building: A Design and Construction Handbook

(NAHBRC, 1994). The document also contains prescriptive span tables for roof 

framing design.
Residential Structural Design Guide 5-71

EXAMPLE 5.8 Ridge Beam Design

Given One-story building

Ridge beam span = 13 ft

Roof slope = 6:12

Rafter horizontal span = 12 ft

Loading (chapter 3)

Dead = 15 psf

Snow = 20 psf

Wind (110 mph, gust) = 6.3 psf (inward)

= 14.2 psf (outward, uplift)

Live = 10 psf

Find Optimum size and grade of lumber to use for a solid (single-member) ridge beam.

Solution

1. Evaluate load combinations applicable to the ridge beam design (see chapter 3, 

table 3.1).

D + (Lr or S) Controls ridge beam design in the inward-bending direction 

(compression side of beam laterally supported by top bearing 

rafters); Lr can be ignored because the roof snow load is 

greater.

0.6 D + 0.6Wu May control ridge beam design in outward-bending direction 

because the bottom (compression side) is laterally unsupported 

(that is, exposed ridge beam for cathedral ceiling); also 

important to ridge beam connection to supporting columns. A 

ridge beam supporting rafters that are tied down to resist wind 

uplift, however, cannot experience significant uplift without 

significant upward movement of the rafters at the wall 

connection and deformation of the entire sloped roof 

diaphragm (depending on roof slope).

D + W Not controlling because snow load is greater in the inward 

direction; also, positive pressure is possible only on the sloped 

windward roof surface, whereas the leeward roof surface is 

always under negative (suction) pressure for wind 

perpendicular to the ridge; wind parallel to the ridge results in 

uplift across both sides of the roof, which is addressed in the 

0.6 D + 0.6Wu load combination and the roof uplift 

coefficients in chapter 3 and based on this worst case wind 

direction.

2. Determine the ridge beam bending load, shear, and moment for the wind uplift load 

case.

In accordance with a procedure similar to step 4 of example 5.7, the 

following ridge beam loads are determined:

Rafter sloped span = horizontal span/cos θ
Residential Structural Design Guide 5-72

= 12 ft/cos 26.6°

= 13.4 ft

Load on ridge beam

wdead = (rafter sloped span)(15 psf)

[1/2 rafter span on each side]

= (13.4 ft)(15 psf)

= 201 plf

0.6 wdead = 121 plf

wwind = (13.4 ft)(14.2 psf) cos 26.6°

= 170 plf

0.6 wwind = 102 plf

wtotal = 102 plf - 121 plf= -19 plf (inward or downward)

Shear, Vmax = 1/2 w = 1/2 (19 plf)(13 ft)

= 124 lb

Moment, Mmax = 1/8 w2 = 1/8 (19 plf)(13ft)2

= 401 ft-lb

Note: If the rafters are adequately tied down to resist uplift from wind, the ridge 

beam cannot deform upward without deforming the entire sloped roof 

diaphragm and the rafter-to-wall connections; therefore, the preceding loads 

should be considered with reasonable judgment. Ensuring that the structure is 

appropriately tied together to act as a unit is more important.

3. Determine the ridge beam loading, shear, and moment for the D + S gravity load 

case.

D + S = 15 psf + 20 psf = 35 psf

(pressures are additive because both are gravity loads)

load on ridge beam

WD+S = (13.4 ft)(35 psf) = 469 plf

Shear, Vmax = 1/2 (469 plf)(13 ft) = 3,049 lb

Moment, Mmax = 1/8 (469 plf)(13 ft)2 = 9,908 ft-lb

4. Determine the optimum ridge beam size and grade based on the preceding bending 

loads and lateral support conditions.

Note: The remainder of the problem is essentially identical to example 5.7 with 

respect to determining the strength of the wood member. A trial member size and 

grade are needed to determine the lumber stresses as well as the lumber property 

adjustment values, however, and the process of optimizing a lumber species, size, 

and grade selection from the multitude of choices is iterative and time consuming 

by hand calculation. Several computerized wood design products on the market can 

perform the task, but those procedures may not allow for flexibility in design 

approach or assumptions if the designer is attempting to use recommendations 

similar to those given in this guide. For this reason, many designers prefer to create 

their own analysis spreadsheets as a customized personal design aid. The remainder 

of this problem is left to the reader for experimentation.
Residential Structural Design Guide 5-73

5.8 References

American Institute of Steel Construction (AISC). 2011. Manual of Steel 

Construction: Allowable Stress Design, 9th Edition. Chicago: American 

Institute of Steel Construction.

American Iron and Steel Institute (AISI). 1993. Residential Steel Beam and 

Column Load/Span Tables, Publication RG-936. Washington, DC: 

American Iron and Steel Institute.

———. 2012. North American Standard for Cold-Formed Steel Framing—

Prescriptive Method for One and Two Family Dwellings, 2007 Edition 

with Supplement 3 (Reaffirmed 2012). AISI S230-07 w/S3-12. 

Washington, DC: American Iron and Steel Institute.

American Society of Agricultural Engineers (ASAE). 1997. Design Requirements 

and Bending Properties for Mechanically Laminated Columns (EP 559). 

St. Joseph, MI: ASAE.

American Society of Civil Engineers (ASCE). 2010. Minimum Design Loads for 

Buildings and Other Structures. ASCE 7-10. Reston, VA: ASCE. 

American Society of Testing and Materials (ASTM). 2007. Standard Practice for 

Establishing Allowable Properties for Visually-Graded Dimension 

Lumber from In-Grade Tests of Full-Size Specimens (D1990-97). West 

Conshohocken, PA: ASTM.

———. 2012. Standard Specification for Computing the Reference Resistance of 

Wood-Based Materials and Structural Connections for Load and 

Resistance Factor Design (D5457-93). West Conshohocken, PA: ASTM.

American Wood Council (AWC). 2012a. Commentary on the National Design 

Specification for Wood Construction. Washington, DC: AWC. 

———. 2012b. National Design Specification for Wood Construction and 

Supplement. Washington, DC: AWC.

———. 2012c. Wood Frame Construction Manual. Washington, DC: AWC.

APA–The Engineered Wood Association (APA). 1998a. Product Design 

Specification (PDS). Tacoma, WA: APA.
Residential Structural Design Guide 5-74

———. 1998b. Product Design Specification (PDS) Supplement. Tacoma, WA. 

APA.

———. 2011a. Engineered Wood Construction Guide. Tacoma, WA: APA.

———. 2011b. Load-Span Tables for APA Structural-Use Panels. Tacoma, WA: 

APA.

Applied Technology Council (ATC). 1999. Minimizing Floor Vibration. ATC 

Design Guide 1. Redwood City, CA: ATC. 

Callahan, E. 2004. Metal Plate Connected Wood Truss Handbook. Madison, WI: 

Wood Truss Council of America.

Douglas, B. K., and P. Line. 1996. “System Effects in Wood Assemblies.” In 

Proceedings of the International Wood Engineering Conference, New 

Orleans.

Forest Products Laboratory. 2010. Wood Engineering Handbook, 2nd edition. 

Englewood Cliffs, NJ: Prentice Hall.

———. 2013. Wood Engineering Handbook, centennial edition. Englewood 

Cliffs, NJ: Prentice Hall.

Forest Products Research Society (FPRS). 1983. “Wall & Floor Systems: Design 

and Performance of Light Frame Structures.” 7317. Madison, WI: FPRS.

Gillespie, R. H., D. Countryman, and R. F. Blomquist. 1978. “Structural Design 

Considerations.” Chapter 3 in Adhesives in Building Construction, 

Agriculture Handbook No. 516. Washington, DC: U.S. Department of 

Agriculture.

U.S. Department of Housing and Urban Development (HUD). 1980. Prevention 

and Control of Decay in Homes. Washington, DC: HUD.

———. 1999. System Performance of Wood Header Assemblies. Prepared for 

HUD by the NAHB Research Center. Washington, DC: HUD.

———. 2015. Durability by Design. Washington, DC: HUD. 

———. 2006. Moisture Resistant Homes. Washington, DC: HUD.

International Code Council (ICC). 2008. Standard for Residential Construction in 

High Wind Regions. ICC-600. Washington, DC: ICC.

———. 2012. International Residential Code. Washington, DC: ICC.
Residential Structural Design Guide 5-75

International Code Council Evaluation Services (ICC-ES). 2013. Power-Driven 

Staples and Nails. ESR-1539. Whittier, CA, ICC-ES.

Mtenga, P. V. 1998. “Impact of Using a System Factor in the Design of LightFrame Roof Assemblies,” Forest Products Journal 48 (5): 27–34.

National Association of Home Builders (NAHB). 1961. Bridging in Residential 

Floor Construction, LR-6. Washington, DC: NAHB.

National Association of Home Builders Research Center (NAHBRC). 1981.

NAHB Beam Series. Washington, DC: NAHB.

———. 1994. Cost Effective Home Building: A Design and Construction 

Handbook. Upper Marlboro, MD: NAHB

———. 1998. Prescriptive Method for Residential Cold-Formed Steel Framing, 

3rd Edition. Upper Marlboro, MD: NAHB.

National Institute of Standards and Technology (NIST). 2009. Voluntary Product 

Standard for Construction and Industrial Plywood (USDOC PS 1). 

Gaithersburg, MD: NIST.

———. 2010a. American Softwood Lumber Voluntary Product Standard

(USDOC PS 20). Gaithersburg, MD: NIST.

———. 2010b. Voluntary Performance Standard for Wood-Based Structural-Use 

Panels (USDOC PS 2). Gaithersburg, MD: NIST.

Pellicane, P. J., and R. W. Anthony. 1996. “Effect of Adhesives on the Deflection 

Behavior of Light-Frame Wood Floors.” In Proceedings of the 

International Wood Engineering Conference, Vol. 3. Madison, WI: 

Omnipress.

Polensek. 1975. Rational Design Procedures for Wood Stud Walls under Bending 

and Compression Loads. Forest Research Laboratory, Oregon State 

University, Corvallis, OR: OSU.

Rosowsky, D., and B. Ellingwood, 1992. “Reliability of Wood Systems Subjected 

to Stochastic Live Loads,” Wood and Fiber Science 24 (1): 47–59.

Truss Plate Institute (TPI). 2007a. National Design Standard for Metal Plate 

Connected Wood Truss Construction (TPI-1-07). Madison, WI: TPI.
Residential Structural Design Guide 5-76

———. 2007b. Commentary and Appendices to the National Design Standard for 

Metal Plate Connected Wood Truss Construction (TPI-1-07). Madison, 

WI: TPI.

Tucker, B. J., and K. J. Fridley. 1999. “Concentrated Load Design Procedures for 

Wood-Joist Floor Systems,” Journal of Structural Engineering 125 (7): 

706–712.

Woeste, L. F., and J. D. Dolan. 1998. “Beyond Code: Preventing Floor 

Vibration,” Journal of Light Construction November: 69–71.

Wolfe, R. W. 1990. “Performance of Light-Frame Redundant Assemblies.” In 

Proceedings of 1990 International Timber Engineering Conference, vol. 

1: 124–131. Madison, WI: Omnipress.

Wolfe, R. W. 1991. “Structural Performance of Light-Frame Truss-Roof 

Assemblies.” Proceedings of the International Wood Engineering 

Conference, vol. 3. Madison, WI: Omnipress.

Wolfe. R.W. 1996. Structural Performance of Light-Frame Truss-Roof 

Assemblies.” Proceedings of the International Wood Engineering 

Conference, vol. 3. Madison, WI: Omnipress.

Wolfe, R. W., and T. LaBissoniere. 1991. Structural Performance of Light-Frame 

Roof Assemblies, II. Conventional Truss Assemblies, FPL-RP-499. 

Madison, WI: USDA Forest Products Laboratory.

Wood Truss Council of America (WTCA). 1999. Commentary for Permanent 

Bracing of Metal Plate Connected Wood Trusses. Madison, WI: WTCA.

———. 2008. Standard Practice for Metal Plate Connected Wood Truss Design 

Responsibilities (WTCA 1-95). Madison, WI: WTCA.
Residential Structural Design Guide 6-1

CHAPTER 6

Lateral Resistance to 

Wind and 

Earthquakes

6.1 General

The objectives in designing a building’s lateral resistance to wind and 

earthquake forces are to protect the health, safety, and welfare of the general 

public by minimizing risk to life that can potentially result from both structural 

and nonstructural damage (FEMA, 2003).

In light-frame construction, the lateral-force-resisting system (LFRS) 

comprises shear walls, diaphragms, and their interconnections to form a wholebuilding system that may behave differently from the sum of its individual parts. 

In fact, shear walls and diaphragms are themselves subassemblies of many parts 

and connections; thus, designing an efficient LFRS system is perhaps the greatest 

challenge in the structural design of light-frame buildings. One key component of 

a viable lateral design is ensuring that the load path is provided from the roof 

down through the walls and floor diaphragms into the foundation and eventually

the soil. This can be more difficult when shear walls are not stacked directly on 

top of one another, necessitating hardware and specific details to drag shear to 

parts of the building that are designed to resist lateral forces. In part, the challenge

results from the lack of any single design methodology or theory that provides

reasonable predictions of complex, large-scale system behavior in conventionally

built or engineered light-frame buildings. At times, current design methods depart 

significantly from observations of system testing or field experience. The desire 

for open floor plans, two-story foyers, great rooms, and large windows common 

in modern homes creates a challenge in locating and designing the elements of the 

LFRS.
Residential Structural Design Guide 6-2

The lateral design of light-frame buildings is not a simple endeavor that

provides “exact” solutions. By the very nature of the LFRS, the real behavior of

light-frame buildings is highly dependent on the performance of building systems,

including the interactions of structural and nonstructural components. For

example, the nonstructural components in conventional housing (that is, sidings,

interior finishes, interior partition walls, and even windows and trim) can account

for more than 50 percent of a building’s lateral resistance. The contribution of

these components is not considered part of the “designed” LFRS, however, for 

lack of appropriate design tools and because building code provisions may 

prohibit such considerations. In addition, the need for simplified design methods 

inevitably leads to a tradeoff: analytical simplicity for design efficiency.

Designer judgment is a crucial factor that comes into play when the 

designer selects how the building is to be analyzed and to what extent the analysis 

is a correct representation of the true design problem. Designer judgment is 

essential in the early stages of design because the analytic methods and 

assumptions used to evaluate the lateral resistance of light-frame buildings are not 

in themselves completely correct representations of the problem.

This chapter focuses on methods for evaluating the lateral resistance of 

individual subassemblies of the LFRS (that is, shear walls and diaphragms) and 

the response of the whole building to lateral loads (that is, load distribution). The 

information and design examples presented in this chapter are intended to provide 

a useful guide and resource to supplement existing building code provisions.

In seismic design, factors that translate into better performance may not 

always be obvious. The designer should become accustomed to thinking in terms 

of the relative stiffness of components that make up the whole building because a 

system-level response often is expected in earthquakes. Important, too, is an 

understanding of the inelastic behavior of wood-framed systems that affects the 

optimization of strength, stiffness, damping, and ductility. Many factors relate to a 

structural system’s deformation capability and the system’s ability to absorb and 

safely dissipate energy from abusive cyclic motion in a seismic event.

Although seismic and wind design of wood-framed residential buildings 

presents some challenges to the designer, significant progress has been made in 

understanding load paths and therefore meeting code objectives with existing 

building codes (Filiatrault et al., 2010) and achieving improved performance 

objectives with alternative methods (van de Lindt et al., 2010). The designer 

should have reasonable knowledge of the underpinnings of current LFRS design 

approaches, including their uncertainties and limitations. Design provisions 

generally are based on an “element-based” approach to engineering and usually 

provide little guidance about how the various elements perform when assembled 

in a real building. The next section presents a brief overview of several wholehouse lateral load tests, whose conclusions are currently being considered by 

relevant code committees.
Residential Structural Design Guide 6-3

6.2 Overview of Whole-Building Tests

A significant number of full-scale tests of houses and even larger woodframed buildings have been conducted to gain insight into actual system strength, 

structural behavior, and system performance. Research prior to the mid-1990s has 

been summarized and is available in reports by Thurston (1994) and NIST (1998).

One whole-house test program investigated the lateral stiffness and natural 

frequency of a production-built home (Yokel, Hsi, and Somes, 1973). The study 

applied a design load simulating a uniform wind pressure of 25 pounds per square 

foot (psf) to a conventionally built home: a two-story, split-foyer dwelling with a 

fairly typical floor plan. The maximum deflection of the building was only 0.04 

inches and the residual deflection about 0.003 inches. The natural frequency and 

dampening of the building were 9 hz (0.11 second natural period) and 6 percent, 

respectively. The testing was nondestructive, such that the investigation yielded 

no information on “post yielding” behavior; however, the performance was good 

for the nominal lateral design loads under consideration.

Another whole-house test applied transverse loads without uplift to a 

wood-framed house. Failure did not occur until the lateral load reached the 

“equivalent” of a 220 mile-per-hour (mph) wind event without inclusion of uplift 

loads (Tuomi and McCutcheon, 1974). The house was fully sheathed with 3/8-

inch plywood panels, and the number of openings was somewhat fewer than 

would be expected for a typical home (at least on the street-facing side). The 

failure took the form of slippage at the floor connection to the foundation sill plate 

(that is, only one 16d toenail at the end of each joist, and the band joist was not 

connected to the sill). The connection was somewhat less than what is now 

required in the United States for conventional residential construction (ICC, 

2012b). The racking stiffness of the walls nearly doubled from that experienced 

before the addition of the roof framing. Also, the simple 2x4 wood trusses were 

able to carry a gravity load of 135 psf—more than three times the design load of 

40 psf. Notably, combined uplift and lateral load, which would be expected in 

high-wind conditions, was not tested. Further, the test house was relatively small 

and “boxy” in comparison to modern homes.

Many whole-house tests have been conducted in Australia. In one series of 

whole-house tests, destructive testing showed that conventional residential 

construction (only slightly different from that in the United States) was able to 

withstand 2.4 times its intended design wind load (corresponding to a 115 mph 

wind speed) without failure of the structure (Reardon and Henderson, 1996). The 

test house had typical openings for a garage, doors, and windows and had no 

special wind-resistant detailing. The tests applied a simultaneous roof uplift load 

of 1.2 times the total lateral load. The drift in the two-story section was 3 

millimeters (mm; approximately 1/8 inch) at the maximum applied load, whereas 

the drift in the open one-story section (that is, an area with no interior walls) was 

3 mm at the design load and 20 mm (approximately 3/4 inch) at the maximum 

applied load.

Again in Australia, a house with fiber cement exterior cladding and 

plasterboard interior finishes was tested to 4.75 times its “design” lateral load 
Residential Structural Design Guide 6-4

capacity (Boughton and Reardon, 1984). The walls were restrained with tie rods 

to resist wind uplift loads, which are required in Australia’s typhoon-prone 

regions. The roof and ceiling diaphragm was found to be stiff; in fact, the 

diaphragm rigidly distributed the lateral loads to the walls. The tests suggested 

that the house had sufficient capacity to resist a wind speed of 65 meters per 

second (145 mph).

Yet another Australian test of a whole house found that the addition of 

interior ceiling finishes reduced the deflection (that is, drift) of one wall line by 75 

percent (Reardon, 1988, 1989). When cornice trim was added to cover or dress 

the wall-ceiling joint, the deflection of the same wall was reduced by another 60 

percent (roughly 16 percent of the original deflection). The tests were conducted 

at relatively low load levels to determine the impact of various nonstructural 

components on load distribution and stiffness.

In the 1990s, several whole-building assembly tests in the United States 

were conducted to develop and validate sophisticated finite-element computer 

models (Kasal, Leichti, and Itani, 1994).

In England, researchers have taken a somewhat different approach by 

moving directly from empirical system data to a simplified design methodology, 

at least for shear walls (Griffiths and Wickens, 1996). This approach applies 

various “system factors” to basic shear wall design values to obtain a value for a 

specific application. System factors account for material effects in various wall 

assemblies, wall configuration effects (that is, number of openings in the wall), 

and interaction effects with the whole building. One factor even accounts for the 

fact that shear loads on wood-framed shear walls in a full brick-veneered building 

are reduced by as much as 45 percent for wind loads—assuming, of course, that 

the brick veneer is properly installed and detailed to resist wind pressures.

A number of whole-building tests have been conducted in Japan and the 

United States by using large-scale shake tables to study the seismic response of 

whole, light-frame buildings. For many of the whole-building tests conducted in 

Japan in the 1990s, the associated reports are available only in Japanese 

(Thurston, 1994). United States-based projects for full-scale testing of light-frame 

wood buildings have increased since 2000. A 2009 report prepared by the 

National Association of Home Builders Research Center (NAHB, 2009) provides 

a good summary of those projects. Filiatrault et al. (2002) tested a rectangular 

two-story house with an integrated one-car garage. The building was full scale,

but overall size was limited to the shake table dimensions; regardless, the test

provided state-of-the-art results. The building was designed in accordance with 

the 1988 Uniform Building Code (ICBO, 1988) and performed well at code level 

and near-fault records from the 1994 Northridge earthquake. Gypsum wall board

(GWB) and stucco were shown to provide a very significant increase in strength 

and stiffness (Filiatrault et al., 2002). As part of the NEESWood Project (van de 

Lindt and Gupta, 2006), Filiatrault et al. (2010) conducted full-scale, triaxial tests 

on a two-story, three-bedroom, 1,800-square-foot townhouse with an integrated 

two-car garage on the twin shake tables of the State University of New York at 

Buffalo (figure 6.1). This building also was designed to the 1988 Uniform 

Building Code (ICBO, 1988). The results showed that for light-frame wood 
Residential Structural Design Guide 6-5

buildings typical of 1990s California, only moderate damage resulted during a 

design-level earthquake, whereas significant and costly damage occurred during 

the maximum credible earthquake (MCE). The test results yielded one key 

conclusion: that little or no risk to life safety was present at the MCE level. The 

NEESWood townhouse test confirmed the earlier conclusion that the GWB and 

stucco added strength and stiffness. Full building results are available in the 

project report by Christovasilis et al. (2007). 

FIGURE 6.1 Twin Garages on Shake Tables, SUNY Buffalo

During the California Universities for Research on Earthquake Effects

(CUREE)-Caltech project, Mosalam and Mahin (2007) tested a three-story 

apartment building with a tuck-under garage. Their conclusions confirmed that 

these types of buildings are prone to torsional response and soft-story collapse.

As part of the NEESWood project, van de Lindt et al. (2010) conducted 

the world’s largest shake table test. Figure 6.2 shows a photograph of the 14,000-

square-foot, six-story apartment building ready for testing at Japan’s E-defense 

facility in Miki, Japan. The building was 40-by-60-foot in plan and 56 ft tall. The 

objectives were to (1) provide a general understanding of how midrise light-frame 

wood buildings perform in a major earthquake, and (2) provide validation for the 

performance-based seismic design philosophy developed within the project.
Residential Structural Design Guide 6-6

Overall performance was excellent at MCE level, but the test structure was 

designed at a level expected to provide seismic performance superior to current 

code (van de Lindt et al., 2013), a method called performance-based design 

(PBD); PBD is mentioned throughout this guide as an alternative method of 

design. Full details about the test are available in the project task report (Pei et al., 

2010).

FIGURE 6.2 Six-Story Building on Shake Table, Miki, Japan

The Insurance Institute for Business and Home Safety has built a full-scale 

wind tunnel to support the research done for their insurance constituency. To date, 

the institute has conducted wind testing on several full-scale houses and 

performed important component testing of siding materials and rooftop 

equipment. Their full-scale facility also has tested products and systems for 

wildfire, hail, and wind-driven rain.

The University of Western Ontario has a full-scale wind testing facility 

called “Three Little Pigs.” That facility has been testing building frames, 

assemblies, and components with some focus on those elements used in lowrise 

buildings.

The growing body of whole-building test data will likely improve 

understanding of the actual performance of light-frame structures in seismic and 

wind events to the extent that the test programs are able to replicate actual 

conditions. Actual performance must also be inferred from anecdotal experience 
Residential Structural Design Guide 6-7

or, preferably, from experimentally designed studies of buildings experiencing 

major seismic or wind events. 

6.3 LFRS Design Steps and Terminology

The LFRS of a home is the “whole house,” including nearly all structural 

and nonstructural components. To enable a rational and tenable design analysis, 

however, the complex structural system of a light-frame house is usually 

subjected to many simplifying assumptions (refer to chapter 2). The following list 

outlines steps required for thoroughly designing a building’s LFRS in typical 

order of consideration.

1. Determine a building’s architectural design, including layout of walls 

and floors (usually predetermined).

2. Design the foundation that will support the load of the structure.

3. Calculate the lateral loads on the structure resulting from wind or 

seismic conditions (refer to chapter 3).

4. Distribute shear loads to the LFRS (wall, floor, and roof systems), 

based on one of the design approaches described later in this chapter 

(refer to section 6.4.1).

5. Determine shear wall and diaphragm assembly requirements for the 

various LFRS components (sheathing thickness, fastening schedule, 

and so forth) to resist the stresses resulting from the applied lateral 

forces (refer to section 6.5).

6. Design the hold-down restraints required to resist overturning forces 

generated by lateral loads applied to the vertical components of the 

LFRS (that is, shear walls).

7. Determine interconnection requirements to transfer shear between the 

LFRS components (that is, roof, walls, floors, and foundation).

8. Evaluate chords and collectors (or drag struts) for adequate capacity 

and for situations requiring special detailing, such as splices.

Depending on the method of distributing shear loads (refer to section 

6.4.1), step 3 may be considered a preliminary design step. If, in fact, loads are 

distributed according to stiffness in step 3, then the LFRS must already be 

defined; therefore, the preceding sequence can become iterative between steps 3 

and 4. A designer need not feel compelled to go to such a level of complexity 

(that is, using a stiffness-based force distribution) in designing a simple home, but 

the decision becomes less intuitive with increasing plan complexity.

The preceding list of design steps introduces several terms that are defined 

herein. 

Horizontal diaphragms are assemblies such as the roof and floors that act 

as “deep beams” by collecting and transferring lateral forces to the shear walls, 
Residential Structural Design Guide 6-8

which are the vertical components of the LFRS. The diaphragm is analogous to a 

horizontal, simply supported beam laid flatwise; a shear wall is analogous to a 

vertical, fixed-end, cantilevered beam. Chapter 2 discussed the function of the 

LFRS and the lateral load path. The reader is referred to that chapter for a 

conceptual overview of the LFRS and to chapter 3 for methodologies to calculate 

lateral loads resulting from wind and earthquake forces. 

Chords are the members (or a system of members) that form a “flange” to 

resist the tension and compression forces generated by the “beam” action of a 

diaphragm or shear wall. As shown in figure 6.3, the chord members in shear 

walls and diaphragms are different members, but they serve the same purpose in 

the beam analogy. A collector or drag strut, which is usually a system of 

members in light-frame buildings, “collects” and transfers loads by tension or 

compression to the shear-resisting segments of a wall line (see figure 6.4a).

In typical light-frame homes, special design of chord members for floor 

diaphragms may involve some modest detailing of splices at the diaphragm 

boundary (that is, joints in the band joists). If adequate connection is made 

between the band joist and the wall top plate, then the diaphragm sheathing, band 

joists, and wall framing function as a “composite” chord to resist the chord forces; 

thus, the diaphragm chord is usually integral with the collectors or drag struts in 

shear walls. Given that the collectors on shear walls often perform a dual role as a 

chord on a floor or roof diaphragm boundary, the designer needs only to verify 

that the two systems are reasonably interconnected along their boundary, thus 

ensuring composite action as well as direct shear transfer (or slip resistance) from 

the diaphragm to the wall. As shown in figure 6.4b, the failure plane of a typical 

“composite” collector or diaphragm chord can involve many members and their 

interconnections.

For shear walls in typical light-frame buildings, tension and compression 

forces on shear wall chords are usually considered. In particular, the connection of 

hold-downs to shear wall chords should be carefully evaluated with respect to the 

transfer of tension forces to the structure below. Tension forces result from the 

overturning action (that is, the overturning moment) caused by the lateral shear 

load on the shear wall. In some cases, a thicker chord may be required to allow for 

an adequate hold-down connection or to withstand the tension and compression 

forces presumed by the beam analogy. Fortunately, most chords in light-frame 

shear walls are located at the ends of walls or adjacent to openings, where 

multiple studs are already required for constructability and gravity load resistance 

(see cross-section B in figure 6.3).
Residential Structural Design Guide 6-9

FIGURE 6.3 Chords in Shear Walls and Horizontal Diaphragms Using

the “Deep Beam” Analogy

WEB (TO RESIST TRANSVERSE SHEAR)
Residential Structural Design Guide 6-10

FIGURE 6.4 Shear Wall Collector and Composite Failure Plane
Residential Structural Design Guide 6-11

Hold-down restraints are devices used to restrain the whole building and 

individual shear wall segments from the overturning that results from the 

overturning moment created by lateral forces. The current engineering approach 

calls for restraints that are typically metal connectors (that is, straps or brackets) 

that attach to and anchor the chords (that is, end studs) of shear wall segments 

(see figure 6.5a). In many typical residential applications, however, overturning 

forces may be resisted by the dead load and the contribution of many component 

connections (see figure 6.5b). Unfortunately (but in reality), this consideration 

may require a more intensive analytic effort and greater degree of designer 

presumption because overturning forces may disperse through many “load paths” 

in a nonlinear fashion. Consequently, the analysis of overturning becomes much 

more complicated; the designer cannot simply assume a single load path through a 

single hold-down connector. Indeed, analytic knowledge of overturning has not 

matured sufficiently to offer an exact, performance-based solution, even though 

experience suggests that the resistance provided by conventional framing has 

proven adequate to prevent collapse in all but the most extreme conditions or 

misapplications (see section 6.2).

Framing and fastenings at wall corner regions are major factors in 

explaining the actual behavior of conventionally built homes, yet no method is 

currently recognized to account for this effect. Several studies have investigated 

corner framing effects in restraining shear walls without using hold-down 

brackets. In one such study, cyclic and monotonic tests of typical 12-ft long 

wood-framed shear walls with 2- and 4-ft corner returns demonstrated that 

overturning forces can be resisted by reasonably detailed corners (that is, 

sheathing fastened to a common corner stud), with the reduction in shear capacity 

only about 10 percent from that realized in tests of walls with hold-downs instead 

of corner returns (Dolan and Heine, 1997c). The corner framing approach can also 

improve ductility (Dolan and Heine, 1997c); that result has been confirmed by 

testing in other countries (Thurston, 1994). Such corner or transverse wall 

detailing is recognized in the conventional IRC wall bracing methods; however, 

engineering design of shear walls is based on the assumption that proper 

anchorage and overturning restraint is provided.
Residential Structural Design Guide 6-12

FIGURE 6.5 Two Types of Hold-Down Restraint and

Basic Analytic Concepts

COMPRESSION FORCE, C, ON

END STUD(S)-COMPRESSION CHORD

FOR LOAD DIRECTION SHOWN

TENSION FORCE, T, ON

END STUD(S)-TENSION CHORD

FOR LOAD DIRECTION SHOWN
Residential Structural Design Guide 6-13

6.4.1 The Current LFRS Design Practice

This section provides a brief overview of the current design practices for 

analyzing the LFRS of light-frame buildings. It highlights the advantages and 

disadvantages of the various approaches but, in the absence of a coherent body of 

evidence, makes no attempt to identify which approach, if any, may be considered 

superior. Where experience from whole-building tests and actual building 

performance in real events permits, the discussion provides a critique of current 

design practices that, for lack of better methods, relies somewhat on an intuitive 

sense for the difference between the structure as it is analyzed and the structure as 

it may actually perform. The intent is not to downplay the importance of 

engineering analysis; rather, the designer should understand the implications of 

the current analytic methods and their inherent assumptions and then put them 

into practice in a suitable manner.

6.4.1 Lateral Force Distribution Methods

The design of the LFRS of light-frame buildings generally follows one of 

three approaches, which are described herein. Each differs in its approach to 

distributing whole-building lateral forces through the horizontal diaphragms to the 

shear walls. Each also varies in the level of calculation, precision, and dependence 

on designer judgment. Although different solutions can be obtained for the same 

design by using the different methods, one approach is not necessarily preferred 

to another. All may be used for the distribution of seismic and wind loads to the 

shear walls in a building; however, some of the most recent building codes may 

place limitations on or display preferences for certain methods.

Tributary Area Approach (Flexible Diaphragm)

The tributary area approach is perhaps the most popular method used to 

distribute lateral building loads. Tributary areas based on building geometry are 

assigned to various components of the LFRS to determine the wind or seismic 

loads on building components (that is, shear walls and diaphragms). The method 

assumes that a diaphragm is relatively flexible in comparison to the shear walls 

such that it distributes forces according to tributary areas rather than according to 

the stiffness of the supporting shear walls. This hypothetical condition is 

analogous to conventional beam theory, which assumes rigid supports (as 

illustrated in figure 6.6 for a continuous horizontal diaphragm (or floor) with three 

supports (that is, shear walls).
Residential Structural Design Guide 6-14

FIGURE 6.6 Lateral Force Distribution by a “Flexible” Diaphragm

(Tributary Area Approach)

In seismic design, tributary areas are associated with uniform area weights 

assigned to the building systems (that is, roof, walls, and floors) that generate the 

inertial seismic load when the building is subject to lateral ground motion (refer to 

chapter 3 on earthquake loads). In wind design, the tributary areas are associated 

NOTE: THE RESULTANT SEISMIC 

SHEAR FORCE ACTS THROUGH THE 

CENTER OF MASS AND THE 

RESULTANT WIND SHEAR FORCE ACTS 

THROUGH THE CENTROID OF THE 

TRIBUTARY SURFACE AREA.
Residential Structural Design Guide 6-15

with the lateral component of the wind load acting on the exterior surfaces of the 

building (refer to chapter 3 on wind loads).

The flexibility of a diaphragm depends on its construction as well as on its 

aspect ratio (length to width). Long, narrow diaphragms, for example, are more 

flexible in bending along the long dimension than are short, wide diaphragms. In 

other words, rectangular diaphragms are relatively stiff in one loading direction 

and relatively flexible in the other. Similarly, long shear walls with few openings 

are stiffer than walls comprising only narrow shear wall segments. Although 

analytic methods are available to calculate the stiffness of shear wall segments 

and diaphragms (refer to section 6.5), the actual stiffness of these systems is 

extremely difficult to predict accurately (refer to section 6.2). If the diaphragm is 

considered infinitely rigid relative to the shear walls, and the shear walls have 

roughly equivalent stiffness, the three shear wall reactions will be roughly 

equivalent (that is, R1 = R2 = R3 = 1/3[w][l]). Were this assumption more accurate, 

the interior shear wall would be overdesigned and the exterior shear walls 

underdesigned using the tributary area method. In many cases, the correct solution 

is probably somewhere between the apparent overdesign and underdesign 

conditions.

The tributary area approach is reasonable when the layout of the shear 

walls is generally symmetrical with respect to even spacing and similar strength 

and stiffness characteristics. The method is particularly appropriate in concept for 

simple buildings with diaphragms supported by two exterior shear wall lines (with 

similar strength and stiffness characteristics) along both major building axes. 

More generally, the major advantages of the tributary area LFRS design method 

are its simplicity and applicability to simple building configurations. In more 

complex applications, the designer should consider possible imbalances in shear

wall stiffness and strength that may cause or rely on torsional response to 

maintain stability under lateral load (see the next section, “Relative Stiffness 

Approach”).

Relative Stiffness Approach (Rigid Diaphragm)

The relative stiffness approach was first contemplated for house design in 

the 1940s. An extensive testing program began, which created a database of 

racking stiffnesses for a multitude of interior and exterior wall constructions used 

in residential construction at that time (NBS, 1948). If the horizontal diaphragm is 

considered stiff relative to the shear walls, then the lateral forces on the building 

are distributed to the shear wall lines according to their relative stiffness. A stiff 

diaphragm may then rotate some degree to distribute loads to all walls in the 

building, not just to walls parallel to an assumed loading direction. Thus, the 

relative stiffness approach considers torsional load distribution as well as 

distribution of the direct shear loads. When torsional force distribution must be 

considered, whether to demonstrate lateral stability of an “unevenly” braced 

building or to satisfy a building code requirement, the relative stiffness design 

approach is the only available option.
Residential Structural Design Guide 6-16

Although the approach is conceptually correct and comparatively more 

rigorous than the tributary area approach, its limitations with respect to reasonably 

determining the real stiffness of shear wall lines (composed of several restrained 

and unrestrained segments and nonstructural components) and diaphragms (also 

affected by nonstructural components and the building plan configuration) render 

its analogy to actual structural behavior uncertain. Ultimately, the method is only 

as good as the assumptions regarding the stiffness of shear walls and diaphragms 

relative to the actual stiffness of a complete building system. As evidenced in the 

previously mentioned whole-building tests and in other authoritative design texts 

on the subject (Ambrose and Vergun, 1987), difficulties in accurately predicting 

the stiffness of shear walls and diaphragms in actual buildings are significant. 

Moreover, unlike the other methods, the relative stiffness design approach is 

iterative in that the distribution of loads to the shear walls requires a preliminary 

design so that relative stiffness may be estimated. One or more adjustments and 

recalculations may be needed before reaching a satisfactory final design.

Analytically considering the effects of stiffness in the distribution of 

lateral forces in an LFRS is instructional, however, even if based on somewhat 

idealized assumptions regarding relative stiffness (that is, the diaphragm is rigid 

over the entire expanse of shear walls). The approach is a reasonable tool when 

the torsional load distribution should be considered in evaluating or demonstrating 

the stability of a building, particularly a building that is likely to undergo 

significant torsional response in a seismic event. Indeed, torsional imbalances 

exist in just about any building and may be responsible for the relatively good 

performance of some light-frame homes when one side (that is, the street-facing 

side of the building) is weaker (that is, less stiff and less strong) than the other 

three sides of the building. This condition is common because of the aesthetic 

desire and functional need for more openings on the front side of a building; 

however, a torsional response in the case of underdesign (that is, “weak” or “soft” 

story) can wreak havoc on a building and constitute a serious threat to life.

Total Shear Approach 

Considered one of the most popular and the simplest of the three LFRS 

design methods, the total shear approach uses the total story shear to determine 

the total amount of shear wall length required on a given story level for each 

orthogonal direction of loading. The amount of shear wall is then “evenly” 

distributed in the story, according to designer judgment. Although the total shear 

approach requires the least amount of computational effort among the three 

methods, it demands good judgment as to the distribution of the shear wall 

elements to address or avoid potential loading or stiffness imbalances. In seismic 

design, loading imbalances may be created when a building’s mass distribution is 

not uniform. In wind design, loading imbalances result when the surface area of 

the building is not uniform (that is, taller walls or steeper roof sections experience 

greater lateral wind load). In both cases, imbalances are created when the center 

of resistance is offset from either the center of mass (seismic design) or the 

resultant force center of the exterior surface pressures (wind design). Thus, the 
Residential Structural Design Guide 6-17

reliability of the total shear approach is highly dependent on the designer’s 

judgment and intuition regarding load distribution and structural response. If used 

indiscriminately without consideration of the factors mentioned, the total shear 

approach to LFRS design can result in poor performance in severe seismic or 

wind events. For small structures such as homes, however, the method has 

produced reasonable designs, especially in view of the overall uncertainty in 

seismic and wind load analysis.

6.4.2 Shear Wall Design Approaches

Once the whole-building lateral loads have been distributed and assigned 

to the floor and roof diaphragms and various designated shear walls, each of these 

subassemblies must be designed to resist the assigned shear loads. As discussed, 

the whole-building shear loads are distributed to various shear walls ultimately in 

accordance with the principle of relative stiffness (whether handled by judgment, 

analytic assumptions per a selected design method, or both). Similarly, the 

distribution of the assigned shear load to the various shear wall segments within a 

given shear wall line is based on the same principle but at a different scale. The 

scale is the subassembly (or shear wall) as opposed to the whole building.

The methods for designing and distributing the forces within a shear wall 

line differ, as described herein. As with the two different approaches described for 

the distribution of lateral building loads, the shear wall design methods place 

different levels of emphasis on analytic rigor and judgment. Ultimately, the 

configuration of the building (that is, whether the walls are inherently broken into 

individual segments by large openings or many offsets in plan dimensions) and 

the required demand (that is, shear load) should drive the choice of a shear wall 

design approach and the resulting construction detailing. Thus, the choice of 

which design method to use is a matter of designer judgment and required 

performance. In turn, the design method itself imposes detailing requirements on 

the final construction, in compliance with the analysis assumptions. Accordingly, 

those decisions affect the efficiency of the design effort and the complexity of the 

resulting construction details.

Segmented Shear Wall Approach

The segmented shear wall (SSW) approach, well recognized as a standard 

design practice, is the most widely used method of shear wall design. It considers 

the shear resisting segments of a given shear wall line as separate “elements,” 

with each segment restrained against overturning by the use of hold-down 

connectors at its ends. Each segment is a fully sheathed portion of the wall 

without any openings for windows or doors. The design shear capacity of each 

segment is determined by multiplying the length of the segment (sometimes 

called segment width) by tabulated unit shear design values, which are available 

in the building codes and newer design standards. In its simplest form, the 
Residential Structural Design Guide 6-18

approach analyzes each shear wall segment for static equilibrium in a manner 

analogous to a cantilevered beam with a fixed end (refer to figures 6.3 and 6.5a). 

In a wall with multiple designated shear wall segments, the typical approach to 

determining an adequate total length of all shear wall segments is to divide the 

design shear load demand on the wall by the unit shear design value of the wall 

construction. The effect of stiffness on the actual shear force distribution to the 

various segments is simply handled by complying with maximum shear wall 

segment aspect ratios required by code (that is, segment height divided by 

segment width). Although an inexact method of handling the problem of shear 

force distribution in a shear wall line, the SSW approach has been in successful 

practice for many years, partly because it uses conservative unit shear design 

values.

The stiffness of a shear wall segment is assumed to be linearly related to 

its length (or its total design shear strength); however, the linear relationship is not 

realistic outside certain limits. For example, stiffness begins to decrease with 

notable nonlinearity once a shear wall segment decreases below a 4-ft length on 

an 8-ft high wall (that is, it has an aspect ratio of 2 or greater). This does not mean 

that wall segments shorter than 4 feet in width cannot be used but rather that the 

designer must consider the effect of relative stiffness in distributing the load. The 

SSW approach is also less favorable when the wall as a system rather than 

individual segments (that is, including sheathed areas above and below openings) 

may be used to economize on design while meeting performance requirements 

(see the section, “Basic Perforated Shear Wall Approach, which follows). 

As shown in figure 6.5, designers commonly either neglect the 

contribution of dead load or assume that the dead load on the wall is uniformly 

distributed, as would be the case under gravity loading only. In fact, unless the 

wall is restrained with an infinitely rigid hold-down device (an impossibility), the 

uniform dead load distribution will be altered as the wall rotates and deflects 

upward during the application of shear force (see figure 6.5b). As a result, 

depending on the rigidity of the framing system above, the dead load will tend to 

concentrate more toward the “high points” in the wall line, as the various 

segments begin to rotate and uplift at their leading edges. Thus, the dead load may 

be somewhat more effective in offsetting the overturning moment on a shear wall 

segment than is suggested by the uniform dead load assumption. Unfortunately, 

this phenomenon involves nonrigid body, nonlinear behavior for which no 

simplified methods of analysis exist. That effect is generally not considered, 

therefore, particularly for walls with specified restraining devices (that is, holddowns) that are, by default, generally assumed to be completely rigid—an 

assumption that is known through testing not to hold true to varying degrees, 

depending on the type of device and its installation.

Basic Perforated Shear Wall Approach

The basic perforated shear wall (PSW) approach is gaining popularity 

among designers. A PSW is a wall that is fully sheathed with wood structural 

panels (that is, oriented strand board or plywood) and that has openings or 
Residential Structural Design Guide 6-19

“perforations” for windows and doors. The ends of the walls—rather than each 

individual segment, as in the SSW approach—are restrained against overturning. 

As for the intermediate segments of the wall, they are restrained by conventional 

or designed framing connections, such as those at the base of the wall that transfer 

the shear force resisted by the wall to the construction below. The capacity of a 

PSW is determined as the ratio of the strength of a wall with openings to the 

strength of a wall of the same length without openings. The ratio is calculated by 

using two empirical equations, which are given in section 6.5. Figure 6.7 

illustrates a PSW.

FIGURE 6.7 Basic Perforated Shear Wall

The PSW design method requires the least amount of special construction 

detailing and analysis among the current shear wall design methods. It has been 

validated in several recent studies in the United States but dates back more than 

20 years to research first conducted in Japan (Dolan and Heine, 1997a, 1997b; 

Dolan and Johnson, 1996a, 1996b; NAHB, 1997, 1998, 1999; Ni et al., 1999; 
Residential Structural Design Guide 6-20

Sugiyama and Matsumoto, 1994). Although it produces the simplest form of an 

engineered shear wall solution, other methods, such as the SSW design method—

all other factors equal—can yield a stronger wall. Conversely, a PSW design with 

increased sheathing fastening can outperform an SSW with more hold-downs but 

weaker sheathing fastening. The point is that for many applications the PSW 

method often provides an adequate and more efficient design; therefore, it should 

be considered an option to the SSW method, as appropriate.

Force Transfer Around Openings in Shear Walls

Wall openings for windows and doors can greatly reduce shear wall 

resistance because of the discontinuity of the load transfers, as well as high force 

concentration around openings. The force transfer around openings (FTAO) 

approach is one of the more recent solutions that designers have been using. This 

method consists of shear walls designed and detailed for FTAO so that nails, 

metal straps, and blocking members may be utilized to transfer loads and 

reinforce corners of openings. In the FTAO method, a rational engineer-based 

analysis is required to estimate the force transfer and choose proper metal 

connectors (Li et al., 2012).

Partially Restrained Shear Wall Design

Modern residential designs frequently create larger, more open spaces than 

home designs have in the past, yet the wall bracing provisions in the codes have 

been largely based on historic practice. The larger, open floor plans provide less 

wall area to resist large lateral forces from either wind or seismic events. Partially 

restrained shear walls consider larger braced wall line spacing, braced wall lines 

that extend beyond the locations of overturning restraints, offsets of braced wall 

lines, and other braced wall line conditions that are not completely consistent with 

a standard SSW approach. Shear strength adjustment factors have been developed 

to account for these braced wall line anomalies and are included in the IRC 

prescriptive braced wall tables. The shear strength adjustment factors for both 

wind and seismic braced walls are provided in section R602 in the IRC (ICC, 

2012b) (Crandell and Martin, 2009).

6.4.3 Basic Diaphragm Design Approach

As described in chapter 2 and earlier in this section, horizontal diaphragms 

are designed by using the analogy of a deep beam laid flatwise. Thus, the shear 

forces in the diaphragm are calculated as for a beam under a uniform load (refer 

to figure 6.6). As is similar to the case of shear walls, the design shear capacity of 

a horizontal diaphragm is determined by multiplying the diaphragm depth (that is, 

depth of the analogous deep beam) by the tabulated unit shear design values 

found in building codes. The chord forces (in the “flange” of the analogous deep 

beam) are calculated as a tension force and compression force on opposite sides 
Residential Structural Design Guide 6-21

of the diaphragm. The two forces form a force couple (that is, moment) that 

resists the bending action of the diaphragm.

To simplify the calculation, the designer commonly assumes that the 

chord forces are resisted by a single chord member that serves as the “flange” of 

the deep beam (that is, a band joist). At the same time, bending forces internal to 

the diaphragm are assumed to be resisted entirely by the boundary member or 

band joist rather than by other members and connections within the diaphragm. In 

addition, other parts of the diaphragm boundary (that is, walls) that also resist the 

bending tension and compressive forces are not considered. Certainly, a vast 

majority of residential roof diaphragms that are not considered “engineered” by 

current diaphragm design standards have exhibited ample capacity in major 

design events. Thus, the beam analogy used to develop an analytic model for the 

design of wood-framed horizontal diaphragms has room for improvement that has 

yet to be explored from an analytic standpoint. The designer should note, 

however, that the wall top plates often are considered the chord and collector 

member in engineering design—so the engineering model is not out of line with 

what is provided in conventional construction.

As with shear walls, openings in the diaphragm affect the diaphragm’s 

capacity; however, no empirical design approach accounts for the effect of 

openings in a horizontal diaphragm as the PSW method does for shear walls. If 

openings are present, therefore, the effective depth of the diaphragm in resisting 

shear forces must either discount the depth of the opening or be designed for shear 

transfer around the opening. If shear forces must be transferred around a large 

opening in a diaphragm, the designer commonly performs a mechanics-based

analysis of the shear transfer around the opening. The analysis is similar to the 

previously described method that uses free-body diagrams for the design of shear 

walls. The reader should refer to other sources for further study of diaphragm 

design (Ambrose and Vergun, 1987; APA, 2007; Diekmann, 1986).

6.5 Design Guidelines 

6.5.1 General Approach

This section outlines methods for designing shear walls (section 6.5.2) and 

diaphragms (section 6.5.3). The two methods of shear wall design are the SSW 

and PSW methods. The selection of a method depends on shear loading demand, 

wall configuration, and the desired simplicity of the final construction. Regardless 

of design method and resulting LFRS, the first consideration is the amount of 

lateral load to be resisted by the arrangement of shear walls and diaphragms in a 

given building. The design loads and basic load combinations that include lateral 

loads in chapter 3, table 3.1, are as follows.

• D + 0.75(0.6W) + 0.75L + 0.75(Lr or S or R) ASD

• D + 0.75(0.7E) + 0.75L + 0.75S ASD

• 0.6D + 0.6W ASD
Residential Structural Design Guide 6-22

• 0.6D + 0.7E ASD

• 0.9D + (1.0W or 1.0E) LRFD

Earthquake load and wind load are considered separately, with shear walls 

designed in accordance with more stringent loading conditions. Even when wind 

governs the design of the LFRS, seismic detailing requirements might still be 

necessary.

Lateral building loads should be distributed to the shear walls on a given 

story by using one of the following methods, as deemed appropriate by the 

designer.

• Tributary area approach.

• Total shear approach.

• Relative stiffness approach.

These methods were described earlier (see section 6.4). In the case of the 

tributary area method, the loads can be immediately assigned to the various shear 

wall lines based on tributary building areas (exterior surface area for wind loads 

and building plan area for seismic loads) for the two orthogonal directions of 

loading (assuming rectangular-shaped buildings and relatively uniform mass 

distribution for seismic design). In the case of the total shear approach, the load is 

considered as a “lump sum” for each story for both orthogonal directions of 

loading. The shear wall construction and total amount of shear wall for each 

direction of loading and each shear wall line are then determined in accordance 

with this section to meet the required load as determined by either the tributary 

area or total shear approach. The designer must be reasonably confident that the 

distribution of the shear walls and their resistance is reasonably “balanced” with 

respect to building geometry and the center of the total resultant shear load on 

each story. As mentioned, both the tributary and total shear approaches have 

produced many serviceable designs for typical residential buildings, provided that 

the designer exercises sound judgment.

In the case of the relative stiffness method, the assignment of loads must 

be based on an assumed relationship describing the relative stiffness of various 

shear wall lines. Generally, the stiffness of a wood-framed shear wall is assumed 

to be directly related to the length of the shear wall segments and the unit shear 

value of the wall construction. The PSW method may assume the relative stiffness 

of various PSW lines to be directly related to the design strength of the various 

PSW lines. Using the principle of moments and a representation of wall racking 

stiffness, the designer can then identify the center of shear resistance for each 

story and determine each story’s torsional load (because of the offset of the load 

center from the center of resistance). Finally, the designer superimposes direct 

shear loads and torsional shear loads to determine the estimated shear loads on 

each of the shear wall lines. 

Design example 6.5 of section 6.6 elaborates on and demonstrates the use 

of the methods of load distribution described. The reader is encouraged to study 
Residential Structural Design Guide 6-23

and critique them. The example contains many concepts and insights that cannot 

be otherwise conveyed without the benefit of a “real” problem.

6.5.2 Shear Wall Design

6.5.2.1 Shear Wall Design Values (Fs)

This section provides unfactored (ultimate) unit shear values for woodframed shear wall constructions that use wood structural panels (WSPs). Other 

wall constructions and framing methods are included as an additional resource. 

The unit shear values given here differ from those in the current codes in that they 

are based explicitly on the ultimate shear capacity as determined through testing. 

The designer should refer to the applicable building code for “code-approved” 

unit shear values, therefore. This guide uses ultimate unit shear capacities as its 

basis to give the designer an explicit measure of the actual capacity and safety 

margin (that is, reserve strength) used in design and to provide for a more 

consistent safety margin across various shear wall construction options. 

Accordingly, it is imperative that the values used in this guide be appropriately 

adjusted in accordance with sections 6.5.2.2 and 6.5.2.3 to ensure an acceptable 

safety margin.

Wood Structural Panels

Table 6.1 provides unit shear values for walls sheathed with WSPs that are 

used according to current building codes. Although the actual capacity of a shear 

wall may not be explicitly known to the designer using the codes’ allowable unit 

shear values, one alleged benefit of using the code-approved design unit shear 

values is that the values are believed to address drift implicitly by way of a 

generally conservative safety margin. Even so, shear wall drift usually is not 

analyzed in residential construction for reasons stated previously.

The values in table 6.1 are based primarily on monotonic tests (that is, 

tests that use single-direction loading). Recently, the effect of cyclic loading on 

wood-framed shear wall capacity has generated considerable controversy. 

Depending on the cyclic test protocol, the resulting unit shear values may be 

above or below those obtained from traditional monotonic shear wall test methods 

(ASTM, 2010, 2012). In fact, realistic cyclic testing protocols and their associated 

interpretations were found to be largely in agreement with the results obtained 

from monotonic testing (Karacabeyli and Ceccotti, 1998). A CUREE testing 

protocol development study (Krawinkler et al., 2000) found that, for WSP walls, 

the backbone curves from cyclic testing reasonably matched monotonic testing. 

Cyclic testing still is very important to understand failures modes and hysteretic 

behavior and is necessary for all shear wall materials. The differences generally 

are in the range of 10 percent (plus or minus) and thus are not significant.

The unit shear values in table 6.1 are based on nailed sheathing 

connections. The use of elastomeric glue to attach WSP sheathing to wood 
Residential Structural Design Guide 6-24

framing members increases the shear capacity of a shear wall by as much as 50 

percent or more (White and Dolan, 1993). Similarly, studies using elastomeric 

construction adhesive manufactured by 3M Corporation have investigated seismic 

performance (that is, cyclic loading) and confirm a stiffness increase of about 65 

percent and a shear capacity increase of about 45 to 70 percent over sheathing 

fastened with nails only (Filiatrault and Foschi, 1991). Rigid adhesives may create 

even greater strength and stiffness increases. The use of adhesives is beneficial in 

resisting shear loads from wind. Panel manufacturers do not recommend gluing 

shear wall panels because of

TABLE 6.1

Nominal Shear Resistance (plf) for Wood Structural Panel 

Shear Walls With Framing of Douglas-Fir, Larch, or 

Southern Pinea, b

Panels Applied Directly to Framing

Nail Spacing at Panel Edges, (inches)

Seismic Wind

Panel 

Grade

Nominal 

Panel 

Thickness 

(inches) 

Minimum 

Nail 

Penetration 

in Framing 

(inches)

(APA, 1998)

Nail Size 

(common 

or 

galvanized

box)

6 4 3 2 6 4 3 2c

5/16 1 1/4 6d 400 600 780 1,020 560 840 1,090 1,430

3/8d 1 3/8 8d 460 720 920 1220 645 1,010 1,290 1,710

Structura

l

7/16d 1 3/8 8d 510 790 1,010 1,340 715 1,105 1,415 1,875

I 15/32 1 3/8 8d 560 860 1,100 1,460 785 1,205 1,540 2,045

15/32 1 1/2 10de 680 1,02

0

1,330 1,740 950 1,430 1,860 2,435

plf = pounds per linear foot. a Values are average nominal unit shear capacity and should be adjusted to determine allowable stress design or load and resistance factor 

design-factored resistances. For other rated panels (not Structural I), the table values should be multiplied by 0.85. b All panel edges should be backed with 2-inch nominal or wider framing. Panels may be installed either horizontally or vertically. Space 

nails at 6 inches on center along intermediate framing members for 3/8-inch panels installed, with the strong axis parallel to studs spaced 

24 inches on center and 12 inches on center for other conditions and panel thicknesses.

c Framing at adjoining panel edges should be 3-inch nominal or wider, and nails should be staggered where nails are spaced 2 inches on 

center. A double thickness of nominal 2-inch framing is a suitable substitute. d The values for 3/8- and 7/16-inch panels applied directly to framing may be increased to the values shown for 15/32-inch panels, 

provided that studs are spaced a maximum of 16 inches on center or the panel is applied with its strong axis across the studs. e Framing at adjoining panel edges should be 3-inch nominal or wider, and nails should be staggered where 10d nails penetrating framing 

by more than 1 5/8 inches are spaced 3 inches or less on center. A double thickness of 2-inch nominal framing is a suitable substitute.

concern with panel buckling that may occur as a result of the interaction of rigid 

restraints with moisture and temperature expansion and contraction of the panels. 

Construction adhesives are routinely used in floor diaphragm construction, 

however, to increase the bending stiffness and strength of floors; in-plane 

(diaphragm) shear is probably affected by an amount similar to that reported for 

shear walls.
Residential Structural Design Guide 6-25

For unit shear values of WSPs applied to cold-formed steel framing, the 

designer should consult Shear Wall Values for Light Weight Steel Framing (AISI, 

1996). Table 6.2 presents some typical unit shear values for cold-formed, steelframed walls with WSP sheathing fastened with #8 screws. Values for powerdriven, knurled pins (similar to deformed shank nails) should be obtained from 

the manufacturer and the applicable code evaluation reports (ICC-ES, 2013).

TABLE 6.2 Nominal Unit Shear Resistance (plf) for Walls With ColdFormed Steel Framing and Wood Structural Panelsa, b

Panel Grade

Panel Type and 

Nominal Thickness 

(inches)

Minimum 

Screw Sizec

Screw Spacing at Panel Edges (inches)d

6 4 3 2

Structural I 7/16 OSB #8 700 915 1,275 1,625

15/32 plywood #8 780 990 1,465 1,700

OSB = oriented strand board. plf = pounds per linear foot.

a Values are average nominal unit shear capacity and should be adjusted to determine allowable stress design or load and resistance factor 

design-factored resistances. b Values apply to 18 gauge (43 mil) and 20 gauge (33 mil) steel C-shaped studs with a 1 5/8-inch flange width and 3 1/2- to 5 1/2-inch 

depth. Studs are spaced a maximum of 24 inches on center.

c The #8 screws should have a head diameter of no less than 0.29 inches, and the screw threads should penetrate the framing so that the 

threads are fully engaged in the steel.

d The spacing of screws in framing members located in the interior of the panels should be no more than 12 inches on center.

Portland Cement Stucco

Ultimate unit shear values for conventional portland cement stucco (PCS) 

wall construction range from 490 to 1,580 pounds per linear foot (plf), based on 

the ASTM E 72 (ASTM, 2010) test protocol and 12 tests conducted by various 

testing laboratories (ICBO, 1969; Testing Engineers, Inc., 1970, 1971). In 

general, nailing the metal lath or wire mesh resulted in ultimate unit shear values 

less than 750 plf, whereas stapling resulted in ultimate unit shear values greater 

than 750 plf. An ultimate design value of 500 plf is recommended unless specific 

details of PCS construction are known. A safety factor of 2 provides a 

conservative allowable design value of about 250 plf. The actual capacity can be 

as much as five times 250 plf, depending on the method of construction, 

particularly the means of fastening the stucco lath material. Past code-approved 

allowable design values are typically about 180 plf, and the AWC seismic and 

wind design supplement (AWC, 2012) lists 360 plf as an ultimate design value. 

One legacy code required the values to be further reduced by 50 percent in higher 

hazard seismic design areas, although the reduction factor may not necessarily 

improve performance with respect to the cracking of the stucco finish in seismic

events (HUD, 1999) (refer to chapter 1 and the discussion in chapter 3 on 

displacement compatibility under seismic load). Using a lower seismic response 

modification factor (R) may be more appropriate than increasing the safety 

margin in a manner that is not explicit to the designer. In fact, an R factor for PCS 

wood-framed walls is not explicitly provided in building codes (perhaps an R of 2 

for wood-framed walls with “shear panels of all other materials” is used). An R 
Residential Structural Design Guide 6-26

factor for PCS walls should probably be in the range of 3 to 4 (without additional 

increases in the safety factor) because the metal lath and its connection to wood 

framing provide some ductility.

The preceding values pertain to PCS that is 7/8-inch thick, with nail or 

staple fasteners spaced 6 inches on center for attaching the metal wire mesh or 

lath to all framing members. Nails are typically 11 gauge by 1 1/2 inches in 

length, and staples typically have 3/4-inch leg and 7/8-inch crown dimensions. 

The unit shear values recommended also apply to stud spacings no greater than 16 

inches on center. Finally, the aspect ratio of stucco wall segments included in a 

design shear analysis should not be greater than 2 (height/width), according to 

past building code practice.

Gypsum Wall Board

Ultimate capacities in testing 1/2-inch thick GWB range from 120 to 320 

plf, depending on the fastening schedule (Patton-Mallory, Gutkowski, and Soltis, 

1984; Wolfe, 1983). Allowable or design unit shear values for GWB sheathing 

range from 75 to 150 plf in current building codes, depending on the construction 

and fastener spacing. At least one legacy building code required the values to be 

reduced by 50 percent in high-hazard seismic design areas (ICBO, 1997). GWB 

certainly is not recommended as the primary seismic bracing for walls, although it 

does contribute to the structural resistance of buildings in all seismic and wind 

conditions. Methods of fastening interior GWB vary and are generally not an 

“inspected” system. Table 6.3 provides estimated ultimate unit shear values for 

GWB sheathing.

TABLE 6.3 Nominal Unit Shear Values (plf) for 1/2-Inch Thick Gypsum 

Wall-Board Sheathinga, b

GWB 

Thickness

Blocking 

Conditionc

Spacing of 

Framing 

(inches)

Fastener Spacing at Pane Edges (inches)

12 8 7 6 4

1/2 inch

Blocked 16 120 210 250 260 320

Unblocked 16 80 170 200 220 250

24 40 120 150 180 220

GWB = gypsum wall-board. plf = pounds per linear foot.

a The values represent average nominal unit shear capacity and should be adjusted to determine allowable stress design or load and 

resistance factor design-factored resistances. b Fasteners should be minimum 1 1/2-inch drywall nails (that is, 5d cooler) or 1-1/4-inch drywall screws (that is, #6 size with bugle head) or 

equivalent, with spacing of fasteners and framing members as shown. c “Blocked” refers to the condition in which all edges of the panels are fastened to framing members; “unblocked” refers to the condition in 

which the panels are placed horizontally, with horizontal joints between the panels not fastened to blocking, or vertically, with the top and 

bottom edges fastened only at stud locations.
Residential Structural Design Guide 6-27

1x4 Wood Let-in Braces and Metal T-braces

Wood let-in braces and metal T-braces are common in conventional 

residential construction and add to the shear capacity of walls. Those braces are 

always used in combination with other wall finish materials that also contribute to 

a wall’s shear capacity. The braces are typically attached to the top and bottom 

plates of walls and at each intermediate stud intersection with two 8d common 

nails. Such braces are not recommended for the primary lateral resistance of 

structures in high-hazard seismic or wind design areas. In particular, values of the 

seismic response modification factor R for walls braced in this manner have not 

been clearly defined for standardized seismic design guidance. Design values are 

available in section R602 of the IRC (ICC, 2012b).

Other Shear-Resisting Wall Facings

Just about any wall facing, finish, or siding material contributes to a wall’s 

shear resistance qualities. Although the total contribution of nonstructural 

materials to a typical residential building’s lateral resistance often is substantial 

(that is, nearly 50 percent if interior partition walls are included), current design 

codes in the United States prohibit considerations of the role of facing, finish, or 

siding. Some suggestions call for a simple and conservative 10-percent increase 

(known as the “whole-building interaction factor”) to the calculated shear 

resistance of the shear walls, or a similar adjustment to account for the added 

resistance and whole-building effects not typically considered in design (Griffiths 

and Wickens, 1996). A comprehensive study by Filiatrault et al. (2010) confirmed 

this effect of GWB and stucco on building stiffness, strength, and seismic 

performance. 

Some other types of wall sheathing materials that provide shear resistance 

include particleboard and fiberboard. Nominal unit shear values for fiberboard 

and particleboard are provided in the National Design Specification for Wood

Construction—Wind and Seismic Supplement (AWC, 2008). The designer should 

consult the building code or manufacturer data for additional information on 

fiberboard and other materials’ shear resistance qualities. In one study that 

conducted tests on various wall assemblies for HUD, fiberboard was not 

recommended for primary shear resistance in high-hazard seismic or wind design 

areas because of potential durability and cyclic loading concerns.

Combining Wall Bracing Materials

When wall bracing materials (that is, sheathing) of the same type are used 

on opposite faces of a wall, the shear values may be considered additive. In highhazard seismic design conditions, dissimilar materials generally are assumed to be 

nonadditive. In wind-loading conditions, dissimilar materials may be considered 

additive for WSPs (exterior) with GWB (interior). A method to add dissimilar 

materials was proposed as part of the Seismic Evaluation and Retrofit of MultiUnit Wood-Frame Buildings with Weak First Stories (FEMA P-807) seismic 
Residential Structural Design Guide 6-28

retrofit methodology for soft-story, wood-framed buildings (FEMA, 2012).

Chapter 6 of the 2012 IRC provides more guidance on combining of wall bracing 

materials in various scenarios: on the same wall or wall line, within the same 

story, and between stories.

When calculating the shear capacity for walls with different facings, the 

designer must take care to apply the appropriate adjustment factors to determine 

the wall construction’s total design racking strength. Most of the adjustment 

factors in the following sections apply only to WSP sheathing; therefore, the 

adjustments in the next section should be made as appropriate before determining 

combined shear resistance.

6.5.2.2 Shear Wall Design Capacity

The unfactored and unadjusted ultimate unit shear resistance values of 

wall assemblies should first be determined in accordance with the guidance 

provided in the previous section for rated facings or structural sheathing materials 

used on each side of the wall. This section provides methods for determining and 

adjusting the design unit shear resistance and the shear capacity of a shear wall by 

using either the PSW or SSW approaches, discussed in section 6.4.2. The design 

approaches and other important considerations are illustrated in the design 

examples of section 6.6.

Perforated Shear Wall Design Approach

The following equations provide the design shear capacity of a PSW.

(units plf) Eq. 6.5-1a

 (units lb) Eq. 6.5-1b 

where,

Fpsw = the design shear capacity (lb) of the perforated shear wall.

Fs = the unfactored (ultimate) and unadjusted unit shear capacity (plf) 

for each facing of the wall construction.

F's = the factored and adjusted design unit shear capacity (plf) for the 

wall construction.

L = the length of the perforated shear wall, which is defined as the 

distance between the restrained ends of the wall line.

1/SF = the safety factor adjustment for use with allowable stress design –

2.0 for wind and 2.5 for seismic design.

φ = the resistance factor adjustment for use with load and resistance 

factor design.
Residential Structural Design Guide 6-29

The PSW method (equations 6.5-1a and b) has the following limits on its 

use—

• The value of Fs for the wall construction should not exceed 1,500 

plf in accordance with section 6.5.2.1. The wall must be fully 

sheathed with WSPs on at least one side. Unit shear values of 

sheathing materials may be combined in accordance with section 

6.5.2.1.

• Full-height wall segments within a PSW should not exceed an 

aspect ratio of 4 (height/width) unless that portion of the wall is 

treated as an opening. (Some codes limit the aspect ratio to 2 or 

3.5, but recent testing—mentioned earlier—has demonstrated 

otherwise.) The first wall segment on either end of a PSW must not 

exceed the aspect ratio limitation.

• The ends of the PSW must be restrained with hold-down devices 

sized in accordance with section 6.5.2.3. Hold-down forces that are 

transferred from the wall above are additive to the hold-down 

forces in the wall below. Alternatively, each wall stud may be 

restrained by using a strap sized to resist an uplift force equivalent 

to the design unit shear resistance F's of the wall.

• Top plates must be continuous, with a minimum connection 

capacity at splices with lap joints of 1,000 pounds (lb) or as 

required by the design condition, whichever is greater.

• Bottom plate connections that transfer shear to the construction 

below (that is, resist slip) should be designed in accordance with 

section 6.5.2.4 and should result in a connection at least equivalent 

to one 1/2-inch anchor bolt at 6 feet on center or two 16d 

pneumatic nails 0.131-inch diameter at 24 inches on center for wall 

constructions with Fs not exceeding 800 plf (ultimate capacity of 

interior and exterior sheathing). Such connections have been 

shown to provide an ultimate shear slip capacity of more than 800 

plf in typical shear wall framing systems (NAHB, 1999); refer to 

section 7.3.6 of chapter 7. 

• Net wind uplift forces from the roof and other tension forces that 

result from structural actions above the wall are transferred 

through the wall by using an independent load path. Wind uplift 

may be resisted with the previously described strapping option, 

provided that the straps are sized to transfer the additional load.

Segmented Shear Wall Design Approach

The following equations are used to determine the adjusted and factored 

shear capacity of a segmented shear wall (SSW):
Residential Structural Design Guide 6-30

Eq. 6.5-2a

Eq. 6.5-2b

where,

Fssw = the design shear capacity (lb) of a single shear wall segment.

Fs = the unfactored (ultimate) and unadjusted unit shear resistance (plf) 

for the wall construction in accordance with section 6.5.2.1 for 

each facing of the wall construction.

F's = the factored (design) and adjusted unit shear resistance (plf) for 

the total wall construction.

Ls = the length of a shear wall segment (total width of the sheathing 

panel[s] in the segment).

1/SF = the safety factor adjustment for use with allowable stress design –

2.0 for wind and 2.5 for seismic design.

φ = the resistance factor adjustment for use with load and resistance 

factor design.

The SSW design method (equations 6.5-2a and b) imposes the following 

limits.

• The aspect ratio of wall segments should not exceed 4 (height/width), 

as determined by the sheathing dimensions on the wall segment. 

(Absent an adjustment for the aspect ratio, current codes may restrict 

the segment aspect ratio to a maximum of 2 or 3.5.).

• The ends of the wall segment should be restrained in accordance with 

section 6.5.2.3. Hold-down forces that are transferred from shear wall 

segments in the wall above are additive to the hold-down forces in the 

wall below.

• Shear transfer at the base of the wall should be determined in 

accordance with section 6.5.2.4. 

• Net wind uplift forces from the roof and other tension forces as a result 

of structural actions described are transferred through the wall using an 

independent load path.

For walls with multiple shear wall segments, the design shear resistance 

for the individual segments may be added to determine the total design shear 

resistance for the SSW line. Alternatively, the combined shear capacity at given 

amounts of drift may be determined by using the load-deformation equations in 

section 6.5.2.4.
Residential Structural Design Guide 6-31

6.5.2.3 Overturning Restraint

Section 6.3 and figure 6.5 address overturning restraint of shear walls in 

conceptual terms. In practice, the two generally recognized approaches to 

providing overturning restraint call for—

• The evaluation of equilibrium of forces on a restrained shear wall 

segment using principles of engineering mechanics.

• The evaluation of unrestrained shear walls considering nonuniform 

dead load distribution at the top of the wall, with restraint provided by 

various connections (sheathing, wall bottom plate, corner framing, and 

so forth).

The first method applies to restrained shear wall segments in both the 

PSW and SSW methods. The first segment on each end of a PSW is restrained in 

one direction of loading; therefore, the overturning forces on that segment are 

analyzed in the same manner as for a SSW. The second method is a valid and 

conceptually realistic method of analyzing the restraint of typical residential wall 

constructions, but it has not yet fully matured. Using basic mechanics, as shown 

in figure 6.8, the following equation for the chord tension and compression forces 

are determined by summing moments around the bottom compression or tension 

side of a restrained shear wall segment.

Eq. 6.5-7a

Eq. 6.5-7b

where,

T = the tension force on the hold-down device (lb).

d = the width of the restrained shear wall segment (ft); for segments 

greater than 4 ft in width, use d = 4 ft.

x = the distance between the hold-down device and the compression 

edge of the restrained shear wall segment (ft); for segments greater 

than 4 ft in width, use x = 4 ft plus or minus the bracket offset 

dimension, if any.

∑ MC = 0

F' (d)(h) T (x) D ( d) (w )(d)( d) 0 2

1

2 D

1 s − − W − =

(F' h D (w )(d)) t x

d T 2 D 1

2 W

1  s − − + 

 



 =

M 0 ∑ T =

(F' h D (w )(d)) c x

d C 2 D 1

2 W

1  s + + + 

 



 =
Residential Structural Design Guide 6-32

F's = the design unit shear capacity (plf) determined in accordance with 

equation 6.5-2a (for both the PSW and SSW methods).

h = the height of the wall (ft).

Dw = the dead load of the shear wall segment (lb); dead load must be 

factored and wind uplift considered in accordance with the load 

combinations of chapter 3.

wD = the uniform dead load supported by the shear wall segment (plf); 

dead load must be factored and wind uplift considered in 

accordance with the load combinations of chapter 3.

t = the tension load transferred through a hold-down device, if any, 

restraining a wall above (lb); if no tension load exists, t = 0.

c = the compression load transferred from wall segments above, if any 

(lb); this load may be distributed by horizontal structural elements 

above the wall (that is, not a concentrated load); if no compression 

load exists, c = 0.

The 4-ft width limit for d and x is imposed on the analysis of overturning 

forces as presented above because with longer shear wall lengths, the contribution 

of the additional dead load cannot be rigidly transferred through deep bending 

action of the wall to have a full effect on the uplift forces occurring at the end of 

the segment, particularly when it is rigidly restrained from uplifting. This effect 

also depends on the stiffness of the construction above the wall that “delivers” 

and distributes the load at the top of the wall. The assumptions necessary to 

include the restraining effects of dead load are no trivial matter and, for that 

reason, designers commonly do not include any beneficial effect of dead load in 

the overturning force analysis of individual shear wall segments.
Residential Structural Design Guide 6-33

FIGURE 6.8 Evaluation of Overturning Forces on a Restrained Shear

Wall Segment

For a more simplified analysis of overturning forces, the effect of dead 

load may be neglected (the normal case in practice) and the chord forces 

determined as follows, using the symbols defined as before.

Eq. 6.5-7c

Any tension or compression force transferred from shear wall overturning 

forces originating above the wall under consideration must be added to the result 

of equation 6.5-7c, as appropriate. Also, any net wind uplift force is assumed to 

be resisted by a separate load path (that is, wind uplift straps are used in addition 

to overturning or hold-down devices).

The proper detailing to balance localized stiffness effects for more even 

force transfer obviously is a matter of designer judgment. It is mentioned here to 

emphasize the importance of detailing in wood-framed construction. In particular, 

wood framing has the innate ability to distribute loads, although weaknesses can 

develop from seemingly insignificant details. The concern noted has been 

attributed to actual problems (that is, bottom plate splitting) only in severe seismic 

events and in relatively heavily loaded shear walls. For that reason, braced wall 

lines now require larger washers on bottom plate anchor bolts, such as a 3-inch 

F' h

x

d T C  s 

 



 = =
Residential Structural Design Guide 6-34

square by 1/4-inch thick plate washer, to prevent the development of cross-grain 

tension forces in bottom plates in high-hazard seismic regions. The development 

of high cross-grain tension stresses poses less concern when nails are used to 

fasten the bottom plate and are located in pairs or staggered on both sides of the 

wood plate; thus, the two connection options mentioned represent different 

approaches. The first, using the plate washers, maintains a rigid connection 

throughout the wall to prevent cross-grain tension in the bottom plate. The 

second, using nails, is a more “flexible” connection that prevents concentrated 

cross-grain bending forces from developing. With sufficient capacity provided, 

the nailing approach may yield a more “ductile” system. 

Finally, the hold-down must be attached to a vertical wall framing member 

(that is, a stud) that receives the WSP edge nailing. If not, the hold-down will not 

be fully effective (that is, the overturning forces must be “delivered” to the holddown through the sheathing panel edge nailing). 

6.5.2.4 Shear Wall Stiffness and Drift

The methods described in this section for predicting shear wall stiffness or 

drift are based on idealized conditions representative solely of the testing 

conditions to which the equations are related. The conditions do not account for 

the many factors that may decrease the actual drift of a shear wall in its final 

construction. As mentioned, shear wall drift is generally overestimated in 

comparison with actual behavior in a completed structure (see section 6.2 on 

whole-building tests). The degree of overprediction may reach a factor of 2 at 

design load conditions. At capacity, the error may not be as large because some 

nonstructural components may be past their yield point.

At the same time, drift analysis may not consider the factors that also 

increase drift, such as deformation characteristics of the hold-down hardware (for 

hardware that is less stiff than that typically used in testing), lumber shrinkage 

(causing time-delayed slack in joints), lumber compression under heavy shear 

wall compression chord load, and construction tolerances. The results of a drift 

analysis should be considered as a guide to engineering judgment, therefore, not 

an exact prediction of drift.

The load-drift equations in this section may be solved to yield shear wall 

resistance for a given amount of shear wall drift. In this manner, a series of shear 

wall segments or even PSWs embedded within a given wall line may be 

combined to determine an overall load-drift relationship for the entire wall line. 

The load-drift relationships are based on the nonlinear behavior of wood-framed 

shear walls and provide a reasonably accurate means of determining the behavior 

of walls of various configurations. The relationships may also be used for 

determining the relative stiffness of shear wall lines in conjunction with the 

relative stiffness method of distributing lateral building loads, and the 

relationships may also be used for considering torsional behavior of a building 

with a nonsymmetrical shear wall layout in stiffness and in geometry. The 

approach is fairly straightforward and is left to the reader for experimentation.
Residential Structural Design Guide 6-35

Perforated Shear Wall Load-Drift Relationship

The load-drift equation that follows is based on several PSW tests already 

discussed in this chapter. The equation provides a nonlinear load-drift relationship 

up to the ultimate capacity of the perforated shear. When considering shear wall 

load-drift behavior in an actual building, the reader is reminded of the 

aforementioned accuracy issues; however, accuracy relative to the test data is 

reasonable (that is, plus or minus 1/2 inch at capacity). 

(inches) Eq. 6.5-8

where

∆ = the shear wall drift (in) at shear load demand, Vd (lb).

G = the specific gravity of framing lumber .

r = the sheathing area ratio.

Vd = the shear load demand (lb) on the perforated shear wall; the 

value of Vd is set at any unit shear demand less than or equal to 

Fpsw,ult, whereas the value of Vd should be set to the design shear 

load when checking drift at design load conditions.

Fpsw,ult = the unfactored (ultimate) shear capacity (lb) for the perforated 

shear wall (that is, Fpsw x SF or Fpsw/φ for ASD and LRFD, 

respectively).

h = the height of wall (ft).

Segmented Shear Wall Load-Drift Relationship

APA Semiempirical Load-Drift Equation

Several codes and industry design guidelines specify a deflection equation 

for shear walls that includes a multipart estimate of various factors’ contribution 

to shear wall deflection (APA, 2007; ICBO, 1997; ICC, 2012a). The approach 

relies on a mix of mechanics-based principles and empirical modifications. The 

principles and modifications are not repeated here because the APA method of 

drift prediction is considered no more reliable than that presented next. In 

addition, the equation is complex relative to the ability to predict drift accurately. 

It also requires adjustment factors, such as a nail-slip factor, that can only be 

determined by testing.

Empirical, Nonlinear Load-Drift Equation

Drift in a WSP shear wall segment may be approximated in accordance 

with the following equation—





 









































 



 ∆ = 8

h

F

V

r

1

G

0.5 1.8

2.8

PSW,ULT

d
Residential Structural Design Guide 6-36

(in) Eq. 6.5-9

where,

∆ = the shear wall drift (in) at load Vd (lb).

G = the specific gravity of framing lumber.

a = the shear wall segment aspect ratio (height/width) for aspect 

ratios from 4 to 1; a value of 1 shall be used for shear wall 

segments with width (length) greater than height.

Vd = the shear load demand (lb) on the wall; the value of Vd is set at 

any unit shear demand less than or equal to Fssw,ult, whereas the 

value of Vd should be set to the design load when checking drift 

at design load conditions.

Fssw,ult = the unfactored (ultimate) shear capacity (lb) of the shear wall 

segment (that is, Fssw x SF or Fssw/φ for allowable stress design 

and load and resistance factor design, respectively).

h = the height of wall (ft).

The preceding equation is based on several tests of shear wall segments 

with aspect ratios ranging from 4:1 to 1:5.

6.5.2.5 Portal Frames

In situations with little space to include sufficient shear walls to meet 

required loading conditions, the designer must turn to alternatives. An example is 

a garage opening supporting a two-story home on a narrow lot such that the 

presence of other wall openings for windows and an entrance door leaves little 

room for shear walls. One option is to consider torsion and the distribution of 

lateral loads in accordance with the relative stiffness method. Another possibility 

is the use of a portal frame.

Portal frames may be simple, specialized framing details that can be 

assembled on site. They use fastening details, metal connector hardware, and 

sheathing to form a wooden moment frame and, in many cases, perform 

adequately. Various configurations of portal frames have undergone testing and 

provide data and details on which the designer can base a design (APA, 2008;

NAHB, 1998). The ultimate shear capacity of portal frames ranges from 2,400 to 

more than 6,000 pounds, depending on the complexity and strength of the 

construction details. A simple detail involves extending a garage header so that it 

is end-nailed to a full-height corner stud, strapping the header to the jamb studs at 

the portal opening, attaching sheathing with a standard nailing schedule, and 

anchoring the portal frame with typical PSW requirements. The system has an 

ultimate shear capacity of about 3,400 pounds that, with a safety factor of 2 to 2.5, 

provides a simple solution for many portal frame applications for residential 

construction in high-hazard seismic or wind regions. Several manufacturers offer 





 



















 



 



 ∆ = 8

h

F

V a G

0.5 2.2

2.8

SSW,ULT

4 d
Residential Structural Design Guide 6-37

pre-engineered portal frame and shear wall elements that can be ordered to 

custom requirements or standard conditions.

6.5.2.6 Moment Frames

Some designs and floor and wall layouts create large open spaces, where 

achieving sufficient lateral load resistance is very difficult with an ordinary LFRS 

system, including portal frames. An example of such an opening is illustrated in 

figure 6.9. This house is located in a hurricane-prone region, where wind speeds 

could be extreme. The front wall is all openings and is relatively narrow, yet 

winds that impact either side of the house will push the house right or left with 

little restraint provided by the front wall. The moment frame in the example is 

steel, which will resist the lateral loads of the wind with minimal deflection. 

The designer must analyze the moment frame as a two-dimensional frame, 

determine the stresses in the members and the connections, and predict the tension 

force that can occur at the moment frame connection to the framing platform. This 

analysis then provides sufficient information to design the steel members and 

specify the connections. Most frames like the one shown in figure 6.9 are shop 

fabricated and field erected. They use steel sections that will fit within the 

confines of wood framing, thus the steel sections will be I- or W-beam shapes or 

rectangular or square steel tubing to which wood framing can easily be attached.

Proprietary moment frames are available that are specified based on loads 

to be resisted and the span and height of the opening. Many of those frames are 

bolted together in the field, but they also are available ready to attach wood 

framing.
Residential Structural Design Guide 6-38

FIGURE 6.9 Moment Frame

Note: Photo courtesy FEMA 55 Coastal Construction Manual.

6.5.3 Diaphragm Design

6.5.3.1 Diaphragm Design Values

Depending on the location and number of supporting shear wall lines, the 

shear and moments on a diaphragm are determined by using the analogy of a 

simply supported or continuous span beam. The designer uses the shear load on 

the diaphragm per unit width of the diaphragm (that is, floor or roof) to select a

combination of sheathing and fastening from a table of allowable horizontal 

diaphragm unit shear values; such tables may be found in U.S. building codes or 

engineering standards. Similar to those for shear walls, unit shear values for 

diaphragms vary according to sheathing thickness and nailing schedules, among 

other factors. Table 6.4 presents several of the more common floor and roof 

constructions used in residential construction, as well as their allowable 

diaphragm resistance values. The values include a safety factor for allowable 
Residential Structural Design Guide 6-39

stress design and therefore require no additional factoring. The aspect ratio of a 

diaphragm should be no greater than 4 (length/width), in accordance with current 

building code limits. In addition, the sheathing attachment in floor diaphragms 

often is supplemented with glue or construction adhesive. The increase in unit 

shear capacity of vertical diaphragms (that is, shear walls) was discussed in 

section 6.5.2.1 in association with table 6.1.

TABLE 6.4

Horizontal Diaphragm Maximum Nominal Shear Values 

(plf) for Unblocked Roof and Floor Construction Using 

Douglas-Fir Framinga, b, c

Panel Type and Application Nominal Panel Thickness 

(inches)

Common Nail 

Size

Design Shear Value (plf)

Structural I (roof)

5/16 6d 460

3/8 8d 670

15/32 10d 800

APA Sturd-I-Floor (floor) and 

rated sheathing

7/16 8d 645

15/32 8d 670

19/32 10d 800

plf = pounds per linear foot.

a Minimum framing member thickness is 1 1/2 inches.

b Nails are spaced at 6 inches on center at supported panel edges and at the perimeter of the diaphragm. Nails spaced at 12 inches on center 

on other framing members are spaced a maximum of 24 inches on center.

c “Unblocked” means that sheathing joints perpendicular to framing members are not fastened to blocking.

Source: AWC NDS Wind and Seismic Supplement 2008

6.5.3.2 Diaphragm Design

As noted, diaphragms are designed in accordance with simple beam 

equations. To determine the shear load on a simply supported diaphragm (that is, 

a diaphragm supported by shear walls at each side), the designer uses the 

following equation to calculate the unit shear force to be resisted by the 

diaphragm sheathing.

Eq. 6.5-10a

Eq. 6.5-10b

where,

Vmax= the maximum shear load on the diaphragm (plf).

w = the tributary uniform load (plf) applied to the diaphragm, resulting 

from seismic or wind loading.

l = the length of the diaphragm perpendicular to the direction of the load 

(ft).

vmax = the unit shear across the diaphragm in the direction of the load (plf).

d = the depth or width of the diaphragm in the direction of the load (ft).

V wl 2

1 max =

d

V v max max =
Residential Structural Design Guide 6-40

The following equations are used to determine the theoretical chord 

tension and compression forces on a simply supported diaphragm, as described 

previously.

Eq. 6.5-11a

Eq. 6.5-11b

where,

Mmax = the bending moment on the diaphragm (ft-lb).

w = the tributary uniform load (plf) applied to the diaphragm, resulting 

from seismic or wind loading.

l = the length of the diaphragm perpendicular to the direction of the load 

(ft).

Tmax = the maximum chord tension force (lb).

Cmax = the maximum chord compression force (lb).

d = the depth or width of the diaphragm in the direction of the load (ft).

If the diaphragm is not simply supported at its ends, the designer uses 

appropriate beam equations (see appendix A) in a manner similar to the preceding 

equation to determine the shear and moment on the diaphragm. The calculations 

to determine the unit shear in the diaphragm and the tension and compression in 

the chords also are similar to those given. The designer should note that the 

maximum chord forces occur at the location of the maximum moment. For a 

simply supported diaphragm, the maximum chord forces occur at mid-span, 

between the perimeter shear walls. Thus, chord requirements may vary depending 

on location and magnitude of the bending moment on the diaphragm. Similarly, 

shear forces on a simply supported diaphragm are highest near the perimeter shear 

walls (that is, reactions).Nailing requirements for diaphragms may therefore be 

adjusted depending on the variation of the shear force in interior regions of the 

diaphragm. Generally, those variations are not critical in small residential 

structures such that fastening schedules can remain constant throughout the entire 

diaphragm. If the horizontal diaphragm contains openings, the width of the 

opening dimension usually is discounted from the width d of the diaphragm when 

determining the unit shear load on the diaphragm.

6.5.3.3 Shear Transfer (Sliding)

The shear forces in the diaphragm must be adequately transferred to the 

supporting shear walls. For typical residential roof diaphragms, conventional roof 

framing connections often are sufficient to transfer the small sliding shear forces 

to the shear walls (unless heavy roof coverings are used in high-hazard seismic 

areas, or steep roof slopes are used in high-hazard wind regions). The transfer of 

2

8

1 Mmax = wl

d

M

T C max max = max =
Residential Structural Design Guide 6-41

shear forces from floor diaphragms to shear walls may also be handled by 

conventional nailed connections between the floor boundary member (that is, a 

band joist or end joist that is attached to the floor diaphragm sheathing) and the 

wall framing below. In heavily loaded conditions, metal shear plates may 

supplement the connections. The simple rule to follow for those connections is 

that the shear force in from the diaphragm must equal the shear force out to the 

supporting wall. Floors supported on a foundation wall usually are connected to a 

wood sill plate bolted to the foundation wall; however, the floor joist or the band 

joist may be directly connected to the foundation wall. Chapter 7 addresses the 

design of these shear connections.

6.5.3.4 Diaphragm Stiffness

Diaphragm stiffness may be calculated using semiempirical methods 

based on principles of mechanics. The equations are found in most modern 

building codes and industry guidelines (APA, 2007; ICC, 2012a). For typical 

residential construction, however, the calculation of diaphragm deflection is 

almost never necessary and is rarely performed; therefore, the equations and their 

empirical adjustment factors are not repeated here. Nonetheless, the designer who 

attempts diaphragm deflection or stiffness calculations is cautioned regarding the 

same accuracy concerns mentioned for shear wall drift calculations. The stiffness 

of floor and roof diaphragms is highly dependent on the final construction, 

including interior finishes (see section 6.2 on whole-building tests).
Residential Structural Design Guide 6-42

6.6 Design Examples

EXAMPLE 6.1 Segmented Shear Wall Design

Given The segmented shear wall line, as shown in the figure below, has the following 

dimensions:

h = 8 ft

L1 = 3 ft

L2 = 2 ft

L3 = 8 ft

Window is 3 ft wide

Wall construction:

• Exterior sheathing is 7/16-inch thick OSB with 8d pneumatic nails (0.113-inch 

diameter by 2.375 inches long) spaced 6 inches on center on panel edges and 

12 inches on center in panel field.

• Interior sheathing is 1/2-inch thick gypsum wall board with #6 screws at 12 

inches on center.

• Framing lumber is Spruce-Pine-Fir, stud grade (specific gravity, G = 0.42);

studs are spaced at 16 inches on center.

Loading condition (assumed for illustration)

Wind shear load on wall line = 3,000 lb

Seismic shear load on wall line = 1,000 lb

Find 1. Design capacity of the segmented shear wall line for wind and seismic shear 

resistance.

2. Base shear connection requirements.

3. Chord tension and compression forces.

4. Load-drift behavior of the segmented shear wall line and estimated drift at 

design.

Solution
Residential Structural Design Guide 6-43

1. Determine the adjusted (design) shear capacities for the wall segments and the total 

wall line (section 6.5.2).

Fs,ext = 715 plf wind OSB sheathing (table 6.1 – OSB is Structural I)

Fs,ext = 510 plf seismic OSB sheathing (table 6.1)

Fs,int = 80 plf GWB sheathing (table 6.3)

The design shear capacity of the wall construction is determined as follows for 

each segment (sections 6.5.2.1 and 6.5.2.2):

F's = F's,ext + F's,int

F's = Fs,ext [1/SF] + Fs,int [1/SF]

SF = 2.0 (wind) or 2.5 (seismic)

Segment 1

a = h/L1 = (8 ft)/(3 ft) = 2.67 (segment aspect ratio)

For wind design

F's,1,wind = (715 plf)(1/2.0) + (80 plf)(1/2.0)

= 358 plf + 40 plf = 398 plf

Fssw,1,wind = F's(L1) = (398 plf)(3 ft) = 1194 lb

For seismic design

F's,1,seismic = (510 plf)(1/2.5) + 0 = 204 plf

Fssw,1,seismic = (204 plf)(3 ft) = 612 lb

Segment 2

For wind design

F's,2,wind = (715 plf)(1/2.0) + (80 plf)(1/2.0)

= 358 plf + 40 plf = 398 plf

Fssw,2,wind = (398 plf)(2 ft) = 796 lb

For seismic design

F's,2,seismic = (510 plf)(1/2.5) + 0 = 204 plf

Fssw,2,seismic = (204 plf)(2 ft) = 408 lb

Segment 3

For wind design

F's,3,wind = (715 plf)(1/2.0) + (80 plf)(1/2.0)

= 358 plf + 40 plf = 398 plf

Fssw,3,wind = (398 plf)(8 ft) = 3184 lb
Residential Structural Design Guide 6-44

For seismic design

F's,3,seismic = (510 plf)(1/2.5) + 0 = 204 plf

Fssw,3,seismic = (204 plf)(8 ft) = 1632 lb

Total for wall line

Fssw,total,wind = 1194 lb + 796 lb + 3184 lb = 5174 lb

Fssw,total,seismic = 612 lb + 408 lb + 1632 lb = 2652 lb

2. Determine base shear connection requirements to transfer shear load to the 

foundation or floor construction below the wall.

The wall bottom plate to the left of the door opening is considered to be continuous 

and therefore acts as a distributor of the shear load resisted by segments 1 and 2. 

The uniform shear connection load on the bottom plate to the left of the opening is 

determined as follows—

Bottom plate length = 3 ft + 3 ft + 2 ft = 8 ft

Base shear resistance required (wind) = (Fssw,1,wind + Fssw,2,wind)/(plate length)

= (1194 lb + 796 lb)/(8 ft) = 249 plf

Base shear resistance required (seismic) = (Fssw,1,seismic + Fssw,2,seismic)/(plate 

length)

= (612 lb + 408 lb)/(8 ft) = 128 plf

For the wall bottom plate to the right of the door opening, the base shear 

connection is equivalent to F's,3,wind = 398 plf or F's,3,seismic = 204 plf for wind and 

seismic design respectively.

Normally, this connection is achieved by use of nailed or bolted bottom plate 

fastenings. Refer to chapter 7 and section 7.3.6 for information on designing these 

connections.

Notes:

1. Although the preceding example shows that variable bottom plate connections 

may be specified based on differing shear transfer requirements for portions of 

the wall, using a constant (that is, worst case) base shear connection to 

simplify construction is acceptable practice; however, doing so can result in 

excessive fastening requirements for certain loading conditions and shear wall 

configurations.

2. For the assumed wind loading of 3,000 lb, the wall has excess design capacity 

(that is, 5174 lb). The design wind load may be distributed to the shear wall 

segments in proportion to their design capacity (as shown in the next step for 

hold-down design) to reduce the shear connection loads accordingly. For 

seismic design, this should not be done, and the base shear connection design 

should be based on the design capacity of the shear walls to ensure that a 

“balanced design” is achieved (that is, the base connection capacity meets or 

exceeds that of the shear wall). That approach is necessary in seismic design 

because the actual shear force realized in the connections may be substantially 

higher than those anticipated by the design seismic load calculated using an R 

factor in accordance with equation 3.8-1 of chapter 3. Refer also to the 

discussion on R factors and overstrength in section 3.8.2 of chapter 3. The 

designer should realize that the GWB interior finish design shear capacity was 
Residential Structural Design Guide 6-45

not included in determining the design shear wall capacity for seismic loading. 

Although this is representative of current building code practice, it can create a 

situation in which the actual shear wall capacity and connection forces 

experienced are higher than those used for design purposes. This condition 

(that is, underestimating of the design shear wall capacity) should also be 

considered in providing sufficiently strong overturning connections (that is, 

hold-downs), as covered in the next step.

3. Determine the chord tension and compression (that is, overturning) forces in the 

shear wall segments (section 6.5.2.3).

Basic equation for overturning (equation 6.5-7c)—

T = C = (d/x)(F's)(h)

Segment 1

h = 8 ft

d = 3 ft

x = d – (width of end studs + offset to center of hold-down anchor bolt)*

 = 3 ft – (4.5 in + 1.5 in)(1ft/12 in) = 2.5 ft

*If an anchor strap is used, the offset dimension may be reduced from that 

determined here, assuming a side-mounted hold-down bracket. Also, depending on 

the number of studs at the end of the wall segment and the type of bracket used, the 

offset dimension will vary and must be verified by the designer.

F's,1,wind = 398 plf

F's,1,seismic = 204 plf

T = C = (3 ft / 2.5 ft)(398 plf)(8 ft) = 3821 lb(wind)

T = C = (3 ft / 2.5 ft)(204 plf)(8 ft) = 1958 lb(seismic)

Segment 2

h = 8 ft

d = 2 ft

x = 2 ft – 0.5 ft = 1.5 ft

F's,2,wind = 398 plf

F's,2,seismic = 204 plf

T = C = (2 ft / 1.5 ft)(398 plf)(8 ft) = 4245 lb(wind)

T = C = (2 ft / 1.5 ft)(204 plf)(8 ft) = 2176 lb(seismic)

Segment 3

h = 8 ft

d = 8 ft

x = 8 ft – 0.5 ft = 7.5 ft

F's,2,wind = 398 plf

F's,2,seismic = 204 plf

T = C = (8 ft / 7.5 ft)(398 plf)(8 ft) = 3396 lb(wind)

T = C = (8 ft / 7.5 ft)(204 plf)(8 ft) = 1741 lb(seismic)
Residential Structural Design Guide 6-46

Notes:

1. In each of the preceding cases, the seismic tension and compression forces on 

the shear wall chords are less than that determined for the wind loading 

condition. This occurrence is the result of using a larger safety factor to 

determine the shear wall design capacity and the practice of not including the 

interior sheathing (GWB) design shear capacity for seismic design. Thus, the 

chord forces based on the seismic shear wall design capacity may be 

underdesigned unless a sufficient safety factor is used in the manufacturer’s 

rated hold-down capacity to compensate. In other words, the ultimate capacity 

of the hold-down connector should be greater than the overturning force that 

could be created based on the ultimate shear capacity of the wall, including the 

contribution of the interior GWB finish. The designer should verify this 

condition because the current code practice may not provide explicit guidance 

on the issue of balanced design on the basis of system capacity (that is, 

connector capacity relative to shear wall capacity). This issue is primarily a 

concern with seismic design because of the higher safety factor used to 

determine design shear wall capacity and the code practice not to include the 

contributing shear capacity of the interior finish.

2. The designer should recognize that the compression chord force is not a point 

load at the top of the stud(s) comprising the compression chord. Rather, the 

compression chord force accumulates through the sheathing, beginning at the 

top of the wall with a value of zero and increasing to C (as determined 

previously) at the base of the compression chord. This condition will affect 

how the compression chord is modeled from the standpoint of determining its 

capacity as a column, using the column equations in the NDS.

3. The design of base shear connections and overturning forces assume that the 

wind uplift forces at the base of the wall are offset by 0.6 times the dead load 

(ASD) at that point in the load path or that an additional load path for uplift is 

provided by metal strapping or other means.

4. As mentioned in step 2 for the design of base shear connections, the wind load 

on the designated shear wall segments may be distributed according to the 

design capacity of each segment in proportion to that of the total shear wall 

line. This method is particularly useful when the design shear capacity of the 

wall line is substantially higher than the shear demand required by the wind 

load, as is applicable to this hypothetical example. Alternatively, a shear wall 

segment may be eliminated from the analysis by not specifying restraining 

devices for the segment (that is, hold-down brackets). If the former approach is 

taken, the wind load is distributed as follows:

Fraction of design wind load to segment 1:

Fssw,1,wind /Fssw,total,wind = (1194 lb)/(5174 lb) = 0.23

Fraction of wind load to segment 2:

Fssw,2,wind /Fssw,total,wind = (796 lb)/(5174 lb) = 0.15

Fraction of wind load to segment 3:

Fssw,3,wind /Fssw,total,wind = (3184 lb)/(5174 lb) = 0.62

Thus, the unit shear load on each shear wall segment resulting from the design 

wind shear of 3,000 lb on the total wall line is determined as follows—

Segment 1: 0.23(3,000 lb)/(3 ft) = 230 plf

Segment 2: 0.15(3,000 lb)/(2 ft) = 225 plf

Segment 3: 0.62(3,000 lb)/(8 ft) = 233 plf
Residential Structural Design Guide 6-47

The overturning forces (chord forces) determined above and the base shear 

connection requirements determined in step 2 may be recalculated by substituting 

the preceding values, which are based on the design wind loading. This approach 

only applies to the wind loading condition when the design wind loading on the 

wall line is less than the design capacity of the wall line. As mentioned, eliminating 

a designed shear wall segment to bring the total design shear capacity more in line 

with the design wind shear load on the wall may be more efficient. Alternatively, a 

lower capacity shear wall construction may be specified to better match the loading 

condition (such as a design using a thinner wood structural sheathing panel). This 

decision will depend on the conditions experienced in other walls of the building 

such that a single wall construction type may be used throughout for all exterior 

walls (that is, simplified construction).

4. Determine the load-drift behavior of the wall line.

Only the load-drift behavior for wind design is shown here. For seismic design, a 

simple substitution of the design shear capacities of the wall segments and the 

safety factor for seismic design (as determined previously) may be used to 

determine a load-drift relationship for use in seismic design.

The basic equation for load-drift estimation of a shear wall segment is as follows—

(equation 6.5-9)

h = 8 ft

G = 0.42 (Spruce-Pine-Fir)

Aspect ratios for the wall segments—

a1 = 2.67

a2 = 4.0

a3 = 1.0

Fssw,ult,1,wind = Fssw,1,wind (SF) = (1196 lb)(2.0) = 2392 lb

Fssw,ult,2,wind = Fssw,2,wind (SF) = (796 lb)(2.0) = 1592 lb

Fssw,ult,3,wind = Fssw,3,wind (SF) = (3184 lb)(2.0) = 6368 lb

...therefore, the total ultimate capacity of the wall for wind loading is

Fssw,ult,wall,wind = 2392 lb + 1592 lb + 6368 lb = 10,352 lb

Substituting the preceding values into the basic load-drift equation yields the 

following load-drift equations for each segment.

Segment 1: ∆1 = 1.03x10-10 (Vd,1,wind)

2.8 (inches)

Segment 2: ∆2 = 2.92x10-10 (Vd,2,wind)

2.8 (inches)

Segment 3: ∆3 = 8.46x10-12 (Vd,3,wind)

2.8 (inches)

Because each segment must deflect equally (or nearly so) as the wall line deflects, 

the above deflections may be set equivalent to the total wall line drift, as follows—





 



















 



 



 ∆ = 8

h

F

V a G

0.5 2.2

2.8

SSW,ULT

4 d
Residential Structural Design Guide 6-48

∆wall = ∆1 = ∆2 = ∆3

Further, the preceding equations may be solved for Vd as follows—

Segment 1: Vd,1,wind = 3683 (∆wall)

0.36

Segment 2: Vd,2,wind = 2541 (∆wall)

0.36

Segment 3: Vd,3,wind = 9003 (∆wall)

0.36

The sum of the preceding equations must equal the wind shear load (demand) on 

the wall at any given drift of the wall, as follows—

Vd,wall,wind = Vd,1,wind + Vd,2,wind + Vd,3,wind = 15,227 (∆wall)

0.36

Solving for ∆wall, the following final equation is obtained for estimating drift and 

any given wind shear load from zero to Fssw,ult,wall,wind.

∆wall = 4.04x10-10(Vd,wall,wind)

2.8

For the design wind load on the wall of 3,000 lb, as assumed in this example, the 

wall drift is determined as follows—

∆wall = 4.04x10-10(3,000)2.8 = 2.2 inches

Note: This analysis, as with most other methods of determining drift, may overlook 

many factors in the as-built construction that serve to increase or decrease drift. As 

discussed in section 6.2, whole-building tests seem to confirm that drift is generally 

overpredicted.

Conclusion In this example, the determination of the design shear capacity of a segmented 

shear wall was presented for seismic design and wind design applications. Issues 

related to connection design for base shear transfer and overturning forces (chord 

tension and compression) were also discussed, and calculations were made to 

estimate these forces using a conventional design approach. In particular, issues 

related to capacity-based design and “balanced design” of connections were 

discussed. Finally, a method to determine the load-drift behavior of a segmented 

shear wall line was presented. The final design may vary based on designer 

decisions and judgments (as well as local code requirements) related to the 

considerations and calculations as given in this example.
Residential Structural Design Guide 6-49

EXAMPLE 6.2 Perforated Shear Wall Design

Given The perforated shear wall, as shown in the following figure, is essentially the same 

wall used in example 6.1. The following dimensions are used—

h = 8 ft

L1 = 3 ft

L2 = 2 ft

L3 = 8 ft

L = 19 ft

A1 = 3.2 ft x 5.2 ft = 16.6 sf (rough window opening area)

A2 = 3.2 ft x 6.8 ft = 21.8 sf (rough door opening area)

Wall construction—

• Exterior sheathing is 7/16-inch thick OSB with 8d pneumatic nails (0.113 

inch diameter by 2 3/8 inches long) spaced 6 inches on center on panel 

edges and 12 inches on center in panel field.

• Interior sheathing is 1/2-inch thick gypsum wall board with #6 screws at 

12 inches on center.

• Framing lumber is Spruce-Pine-Fir, stud grade (specific gravity, G = 

0.42); studs are spaced at 16 inches on center.

Loading condition (assumed for illustration):

Wind shear load on wall line = 3,000 lb

Seismic shear load on wall line = 1,000 lb

Find 1. Design capacity of the perforated shear wall line for wind and seismic shear 

resistance.

2. Base shear connection requirements.

3. Chord tension and compression forces.

4. Load-drift behavior of the perforated shear wall line and estimated drift at 

design load conditions.

Solution

1. Determine the adjusted (design) shear capacity for the perforated shear wall line.

F's = Fs [1/SF] (Eq. 6.5-1a)

SF = 2.0 (wind design) or 2.5 (seismic design)
Residential Structural Design Guide 6-50

Fs = Fs,ext + Fs,int (section 6.5.2.1)

Fs,ext = 715 plf (table 6.1)

Fs,int = 80 plf (table 6.3)

For wind design—

Fs,wind = 715 plf + 80 plf = 795 plf

F's,wind = (795 plf)(1/2.0) = 398 plf

For seismic design—

Fs,seismic = 510 plf + 0 plf = 510 plf

F's,seismic = (510 plf(1/2.5) = 204 plf

The design capacity of the perforated shear wall is now determined as follows—

Fpsw = F's L (Eq. 6.5-1b)

Assume for the sake of this example that the roof dead load supported at the top of 

the wall is 225 plf and that the design wind uplift force on the top of the wall is 

0.6(225 plf) – 400 plf = -265 plf (net design uplift). Thus, for wind design in this 

case, no dead load can be considered on the wall.

wD = 0.6*(225 plf) = 135 plf

*The 0.6 factor comes from the load combinations 0.6D + (W or 0.7E) or 0.6D –

Wu as given in chapter 3.

For wind design,

Fpsw,wind = (398 plf)(19 ft) = 7562 lb

For seismic design,

Fpsw,seismic = (204 plf)(19 ft) = 3876 lb

Note: In example 6.1, using the segmented shear wall approach, the design shear 

capacity of the wall line was estimated as 5174 lb (wind) and 2652 lb (seismic) 

when all of the segments were restrained against overturning by use of hold-down 

devices. Given that the design shear load on the wall is 3,000 lb (wind) and 1,000 

lb (seismic), however, the perforated shear wall design capacity as determined 

above is adequate, and significantly higher than that of the segmented shear wall. 

Hold-downs are therefore required only at the wall ends (see step 3).

2. Determine the base shear connection requirement for the perforated shear wall.

If the wall has a continuous bottom plate that serves as a distributor of the shear 

forces resisted by various portions of the wall, the base shear connection could be 

based on the perforated shear wall’s design capacity as determined in step 1, as 

follows—

For wind design,

UNIFORM BASE SHEAR = (7562 LB)/19 FT = 398 PLF
Residential Structural Design Guide 6-51

For seismic design,

UNIFORM BASE SHEAR = (3876 LB)/19 FT = 204 PLF

The wall bottom plate is not continuous in this example, however; therefore, the 

base shears experienced by the portions of the wall to the left and right of the door 

opening are different, as was the case in the segmented shear wall design approach 

of example 6.1. As a conservative solution, the base shear connection could be 

designed to resist the design unit shear capacity of the wall construction, F's,wind = 

398 plf or F's,seismic = 204 plf. Newer codes that recognize the perforated shear 

method may require this more conservative approach to be used when the bottom 

plate is not continuous such that it serves as a distributor (that is, similar in 

function to a shear wall collector except shear transfer is out of the wall instead of 

into the wall). Of course, the bottom plate must be continuous, and any splices 

must be adequately detailed in a fashion similar to collectors (see example 6.3).

Testing has shown that for walls constructed similar to the one illustrated in this 

example, a bottom plate connection of two 16d pneumatic nails (0.131-inch 

diameter by 3 inches long) at 16 inches on center or 5/8-inch diameter anchor bolts 

at 6 feet on center provides suitable shear and uplift resistance—at least equivalent 

to the capacity of the shear wall construction under conditions of no dead load or 

wind uplift (NAHB, 1999). For other conditions, that connection must be designed 

following the procedures given in chapter 7, using the conservative assumptions as 

stated above.

As an alternative base connection that eliminates the need for hold-down brackets 

at the ends of the perforated shear wall, straps can be fastened to the individual 

studs to resist the required uplift force of 398 plf, as applicable to this example. If 

the studs are spaced 16 inches on center, the design capacity of the strap must be 

(398 plf)(1.33 ft/stud) = 529 lb per stud. If an uplift load due to wind uplift on the 

roof must also be transferred through these straps, the strap design capacity must 

be increased accordingly. In this example, the net wind uplift at the top of the wall 

was assumed to be 265 plf. At the base of the wall, the uplift is 265 plf – 0.6(8 

ft)(8 psf) = 227 plf. Thus, the total design uplift restraint must provide 398 plf + 

227 plf = 625 plf. On a per-stud basis (16 inch on center framing), the design load 

is 1.33 ft/stud x 625 plf = 831 lb/stud. That value must be increased for studs 

adjacent to wall openings, where the wind uplift force is increased, and can be 

achieved by using multiple straps or by specifying a larger strap in those locations. 

Of course, the above combination of uplift loads assumes that the design wind 

uplift load on the roof occurs simultaneously with the design shear load on the 

wall; however, this condition is usually not representative of actual conditions, 

depending on wind orientation, building configuration, and the shear wall location 

relative to the uplift load paths.

3. Determine the chord tension and compression forces.

Left end of the wall (segment 1 in example 6.1)—

T = 3821 lb (wind design)

T = 1958 lb (seismic design)

Right end of the wall (segment 3 in example 6.1)—
Residential Structural Design Guide 6-52

T = 3396 lb (wind design)

T = 1741 lb (seismic design)

Note: One tension bracket (hold-down) is required at each the end of the 

perforated shear wall line but not on the interior segments. Also, refer to the notes 

in example 6.1 regarding “balanced design” of overturning connections and base 

shear connections, particularly when designing for seismic loads.

4. Determine the load-drift behavior of the perforated shear wall line.

The basic equation for load-drift estimation of a perforated shear wall line is as 

follows (section 6.5.2.6)—

(Eq. 6.5-8)

h = 8 ft

G = 0.42 (specific gravity for Spruce-Pine-Fir)

r = 1/(1+α/β)

α = ΣAo/(h x L) = (A1 + A2)/(h x L) 

= (16.6 sf + 21.8 sf)/(8 ft)(19 ft) = 0.25

β = ΣLi/L = (L1 + L2 + L3)/L 

= (3 ft + 2 ft + 8 ft)/(19 ft) = 0.68

r = 1/(1+0.25/0.68) = 0.73

Fpsw,ult,wind = (Fpsw,wind)(SF) = (7562 lb)(2.0) = 15,124 lb

Fpsw,ult,seismic = (Fpsw,seismic)(SF) = (3876 lb)(2.5) = 9690 lb

Substituting in the above equation,

∆wind = 7.9x10-13(Vd,wind)

2.8

∆seismic = 2.7x10-12(Vd,seismic)

2.8

For the design wind load of 3,000 lb and the design seismic load of 1,000 lb 

(assumed for the purpose of this example), the drift is estimated as follows—

∆wind = 7.9x10-13(3,000)2.8 = 0.004 inch

∆seismic = 2.7x10-12(1,000)2.8 = 0.01 inch

Note: The reader is reminded of the uncertainties in determining drift, as discussed 

in example 6.1 and also in chapter 6. For seismic design, some codes may require 

the design seismic drift to be amplified (multiplied) by a factor of 4 to account for 

the potential actual forces that may be experienced relative to the design forces 

that are determined using an R factor; refer to chapter 3 for additional discussion. 

Thus, the amplified drift may be determined as 4 x 0.01 inch = 0.04 inch. If the 

seismic shear load is magnified, however (that is, 4 x 1,000 lb = 4,000 lb), to 

account for a possible actual seismic load (not modified for the seismic response 

of the shear wall system), the seismic drift calculated in the above equation 

becomes 0.03 inch, which is minimally greater. The load adjustment is equivalent 

to the use of an R of 1.5 instead of 6 in equation 3.8-1 of chapter 3. This latter 

approach of magnifying the load is not required in the existing building codes for 

drift determination, however. As mentioned, drift is not usually considered in 

residential design. Finally, the preceding equations may be used to determine a 





 









































 



 ∆ = 8

h

F

V

r

1

G

0.5 1.8

2.8

PSW,ULT

d
Residential Structural Design Guide 6-53

load-drift curve for a perforated shear wall for values of Vd ranging from 0 to 

Fpsw,ult. Although the curve represents the nonlinear behavior of a perforated shear 

wall, it should only be considered a representation, not an exact solution.

Conclusion In this example, the determination of the design shear capacity of a perforated 

shear wall was presented for seismic design and wind design applications. Issues 

related to connection design for base shear transfer and overturning forces (chord 

tension) were also discussed, and calculations (or conservative assumptions) were 

made to estimate these forces. In particular, issues related to capacity-based design 

and “balanced design” of connections were discussed. Finally, a method to 

determine the load-drift behavior of a perforated shear wall line was presented. 

The final design may vary, based on designer decisions and judgments (as well as 

local code requirements) related to the considerations and calculations as given in 

this example.

The chord tension and compression forces are determined following the same 

method as used in example 6.1 for the segmented shear wall design method, but 

only for the first wall segment in the perforated shear wall line (that is, the 

restrained segment); therefore, the tension forces at the end of the wall are 

identical to those calculated in example 6.1.
Residential Structural Design Guide 6-54

EXAMPLE 6.3 Shear Wall Collector Design

Given The example shear wall, assumed loading conditions, and dimensions are shown in 

the figure below.

Find The maximum collector tension force.

Solution The collector force diagram that follows is based on the shear wall and loading 

conditions in the preceding figure.

The first point at the interior end of the left shear wall segment is determined as 

follows—

200 plf (3 ft) – 333 plf (3 ft) = - 400 lb (compression force)

The second point at the interior end of the right shear wall segment is determined 

as follows—

- 400 lb + 200 plf (9 ft) = 1,400 lb (tension force)

The collector load at the rightmost end of the wall returns to zero, as follows—

1,400 lb – 375 plf (8 ft) + 200 plf (8 ft) = 0 lb

Conclusion The maximum theoretical collector tension force is 1,400 lb at the interior edge of 
Residential Structural Design Guide 6-55

the 8-ft shear wall segment. The analysis does not consider the contribution of the 

“unrestrained” wall portions that are not designated shear wall segments and that 

would serve to reduce the amount of tension (or compression) force developed in 

the collector. In addition, the load path assumed in the collector does not consider 

the system of connections and components that may share load with the collector 

(wall sheathing and connections, floor or roof construction above and their 

connections, and so forth). The collector load determined by assuming the top plate 

acts as an independent element can be considered very conservative, therefore, 

depending on the wall-floor/roof construction conditions. Regardless, typical 

practice is to design the collector (and any splices in the collector) to resist a 

tension force as calculated in this example. The maximum compressive force in the 

example collector is determined by reversing the loading direction and is equal in 

magnitude to the maximum tension force. Compressive forces rarely are a concern 

when at least a double top plate is used as a collector, particularly when the 

collector is braced against lateral buckling by attachment to other construction (as 

would generally be necessary to deliver the load to the collector from elsewhere in 

the building).
Residential Structural Design Guide 6-56

EXAMPLE 6.4 Horizontal (Floor) Diaphragm Design

Given The example floor diaphragm and its loading and support conditions are shown in 

the figure below. The relevant dimensions and loads are as follows—

 d = 24 ft

 l = 48 ft

 w = 200 plf (from wind or seismic lateral load)*

*Related to the diaphragm’s tributary load area; see chapter 3 and discussions in 

this chapter.

The shear walls are equally spaced, and it is assumed that the diaphragm is flexible 

(that is, experiences beam action) and that the shear wall supports are rigid. This 

assumption is not correct because the diaphragm may act as a “deep beam” and 

distribute loads to the shear wall by “arching” action rather than bending action. 

Also, the shear walls cannot be considered to be perfectly rigid or to exhibit 

equivalent stiffness except when designed exactly the same with the same 

interconnection stiffness and base support stiffness. Regardless, the assumptions 

made in this example are representative of typical practice.

Find 1. The maximum design unit shear force in the diaphragm (assuming simple 

beam action) and the required diaphragm construction.

2. The maximum design moment in the diaphragm (assuming simple beam 

action) and the associated chord forces.

Solution

1. The maximum shear force in the diaphragm occurs at the center shear wall support. 

Using the beam equations in appendix A for a 2-span beam, the maximum shear 

force is determined as follows—

The maximum design unit shear in the diaphragm is determined as follows—

From table 6.4, the lightest unblocked diaphragm provides adequate resistance. 

Unblocked means that the panel edges perpendicular to the framing (that is, joists 

3,000 lb

2

48 ft (200 plf) 8

5

2

l

w

8

5 Vmax  =



 



  =



 



 =
Residential Structural Design Guide 6-57

or rafters) are not attached to blocking. The perimeter, however, is attached to a 

continuous member to resist chord forces. For typical residential floor construction, 

a 19/32-inch thick subfloor may be used, which would provide at least 800 plf of 

design shear capacity. In typical roof construction, a minimum 15/32-inch thick 

sheathing is used, which would provide about 800 plf of design shear capacity; 

however, residential roof construction usually does not provide the edge conditions 

(that is, a continuous band joist of 2x lumber) associated with the diaphragm values 

in table 6.4 Regardless, roof diaphragm performance has rarely (if ever) been a 

problem in light-frame residential construction, and these values often are used to 

approximate roof diaphragm design values.

Note: The shear forces at other regions of the diaphragm and at the locations of the 

end shear wall supports can be determined in a similar manner using the beam 

equations in appendix A. These shear forces are equivalent to the connection forces 

that must transfer shear between the diaphragm and the shear walls at the ends of 

the diaphragm. For the center shear wall, however, the reaction (connection) force 

is twice the unit shear force in the diaphragm at that location (see beam equations 

in appendix A); therefore, the connection between the center shear wall and the 

diaphragm in this example must resist a design shear load of 2 x 125 plf = 250 plf. 

This load is very dependent on the assumption of a “flexible” diaphragm and 

“rigid” shear walls, however.

2. The maximum moment in the diaphragm also occurs at the center shear wall 

support. Using the beam equations in appendix A, the maximum moment is 

determined as follows—

The maximum chord tension and compression forces are at the same location and 

are determined as follows, based on the principle of a force couple that is 

equivalent to the moment.

The chord members (that is, band joist and associated wall or foundation framing 

that is attached to the chord) and splices must therefore be able to resist 600 lb of 

tension or compression force. Generally, these forces are adequately resisted by the 

framing systems bounding the diaphragm; however, the adequacy of the chords 

should be verified by the designer, based on experience and analysis as above.

Conclusion In this example, the basic procedure and principles for horizontal diaphragm design 

were presented. Assumptions required to conduct a diaphragm analysis based on 

conventional beam theory were also discussed.

14,400ft lb

2

48ft (200plf) 8

1

2

l

w

8

1 M

2 2

max  = − 

 



  =  



 =
Residential Structural Design Guide 6-58

EXAMPLE 6.5 Horizontal Shear Load Distribution Methods

Given General

In this example, the first floor plan of a typical two-story house with an 

attached garage (see the following figure) is used to demonstrate the three 

methods of distributing shear loads discussed in chapter 6, section 6.4.2. The 

first story height is 8 ft (that is, 8 ft ceiling height). Only the load in the NorthSouth (N-S) direction is considered in the example; in a complete design, the 

load in the East-West (E-W) direction would also need to be considered.

Lateral Load Conditions

The following design N-S lateral loads are determined for the story under 

consideration, using the methods described in chapter 3 for wind and seismic 

loads. A fairly high wind load and seismic load condition is assumed for the 

purpose of the example.

Design N-S Wind Lateral Load (120 mph gust, exposure B)
Residential Structural Design Guide 6-59

House: 17,411 lb total story shear

Garage: 3,928 lb total story shear

Total: 21,339 lb

Design N-S Seismic Lateral Load (mapped Ss = 1.5g)

House: 7,493 lb total story shear (tributary weight is 37,464 lb)

Garage: 1,490 lb total story shear (tributary weight is 7,452 lb)

Total: 8,983 lb

Designation of Shear Walls in N-S Direction

Initially, four N-S lines are designated in the first story for shear wall 

construction. The wall lines are A, B, D, and E. If needed, an interior wall line 

may also be designated and designed as a shear wall (see wall line C in the 

preceding figure).

The available length of full-height wall segments in each N-S shear wall line is 

estimated as follows from the floor plan

Wall Line A: 2 ft + 2 ft = 4 ft (garage return walls)

Wall Line B: 1.33 ft* + 11 ft + 9 ft = 20 ft (garage/house shared wall)

Wall Line D: 14 ft = 14 ft (den exterior wall) 

Wall Line E: 2 ft + 3 ft + 2 ft = 7 ft (living room exterior wall)

Total: = 45 ft

*The narrow 1.33 ft segment is not included in the analysis because the 

segment’s aspect ratio is 8 ft/1.33 ft = 6 ft, which is greater than the maximum 

allowable of 4 ft. Some current building codes may restrict the segment aspect 

ratio to a maximum of 2 or 3.5, depending on the code and the edition in local 

use. In such a case, many of the usable shear wall segments would be 

eliminated (that is, all of the 2 ft segments). Thus, the garage opening wall 

would require larger segments, a portal frame (see section 6.5.2.5), or transfer 

of the garage shear load to the house by torsion (that is, treat the garage as a 

cantilever projecting from the house under a uniform lateral load).

Find 1. Using the “total shear method” of horizontal shear load distribution, 

determine the total length of shear wall required and the required shear 

wall construction in the N-S direction.

2. Using the “relative stiffness method” of horizontal shear load distribution, 

determine the shear loads on the N-S shear wall lines.

Solution

1. Using the total shear approach, determine the unit shear capacity required, 

based on the given amount of available shear wall segments in each N-S wall 

line and the total N-S shear load.

In this part of the example, it is assumed that the wall lines will be designed as 

segmented shear wall lines. From the given information, the total length of N-S 

shear wall available is 45 ft. Typical practice in this method is to not include 

segments with aspect ratios greater than 2 because stiffness effects on the 

narrow segments are not explicitly considered. This would eliminate the 2 ft 

segments, and the total available length of shear wall would be 45 ft – 8 ft = 37 

ft in the N-S direction.

The required design unit shear capacity of the shear wall construction and 

ultimate capacity are determined as follows for the N-S lateral design loads.
Residential Structural Design Guide 6-60

Wind N-S

F's,wind = (21,339 lb)/37 ft = 576 plf

Fs, wind = (F's,wind)(SF) = (576 plf)(2.0) = 1,152 plf

Thus, the unfactored (ultimate) and unadjusted unit shear capacity for the shear 

walls must meet or exceed 1,152 plf. Assuming that standard 1/2-thick GWB 

finish is used on the interior wall surfaces (80 plf minimum from table 6.3), the 

required ultimate capacity of the exterior sheathing is determined as follows—

Fs,wind = Fs,ext + Fs,int 

Fs,ext = 1,152 plf – 80 plf = 1,072 plf

From table 6.1, any of the wall constructions that use 2-inch nail spacing at the 

panel perimeter exceed this requirement. By specifying a 3/8-inch thick 

Structural I wood structural panel with 8d common nails spaced at 2 inches on 

center on the panel edges (12 inches on center in the panel field), the design of 

the wall construction is complete, and hold-down and base shear connections 

must be designed.

Seismic N-S

F's,seismic = (8,983 lb)/37 ft = 243 plf

Fs, seismic = (F's,seismic)(SF) = (243 plf)(2.5) = 608 plf

Thus, the unfactored (ultimate) and unadjusted unit shear capacity for the wall 

line must meet or exceed 608 plf. Because seismic codes do not permit the 

consideration of a 1/2-thick GWB interior finish, the required ultimate capacity 

of the exterior sheathing is determined as follows:

Fs,seismic = Fs,ext = 608 plf 

From table 6.1, any of the wood structural panel wall constructions that use 6-

inch nail spacing at the panel perimeter exceed this requirement. By specifying 

3/8-inch thick Structural I wood structural panels with 8d common nails 

spaced at 6 inches on center on the panel edges (12 inches on center in the 

panel field), the design of the wall construction is complete, and hold-down

and base shear connections must be designed.

The base shear connections may be designed in this method by considering the 

total length of continuous bottom plate in the N-S shear wall lines. As 

estimated from the plan, this length is approximately 56 feet. Thus, the base 

connection design shear load (parallel to the grain of the bottom plate) is 

determined as follows:

Base wind design shear load = (21,339 lb)/(56 ft) = 381 plf

Base seismic design shear load = (8,983 lb)/(56 ft) = 160 plf

The base shear connections may be designed and specified following the 

methods discussed in chapter 7—Connections. A typical 5/8-inch diameter 

anchor bolt spaced at 6 feet on center or standard bottom plate nailing may be 

able to resist as much as 800 plf (ultimate shear capacity), which would 

provide a “balanced” design capacity of 400 plf or 320 plf for wind and 

seismic design, with safety factors of 2.0 and 2.5, respectively. Thus, a 

conventional wall bottom plate connection may be adequate for the preceding 

conditions; refer to chapter 7 for connection design information and the 
Residential Structural Design Guide 6-61

discussion in section 7.3.6 for more details on tested bottom plate connections.

If the roof uplift load is not completely offset by 0.6 times the dead load at the 

base of the first story wall, then strapping to transfer the net uplift from the 

base of the wall to the foundation or construction below must be provided.

The hold-down connections for each shear wall segment in the designated 

shear wall lines are designed in the manner shown in example 6.1. Any 

overturning forces originating from shear walls on the second story must also 

be included, as described in section 6.4.2.4.

Notes:

1. The contribution of the interior walls to the lateral resistance is neglected 

in the preceding analysis for wind and seismic loading. As discussed in 

chapter 6, these walls can contribute significantly to the lateral resistance 

of a home and serve to reduce the designated shear wall loads and 

connection loads through alternate, “nondesigned” load paths. In this 

example, approximately 40 ft of interior partition walls in the N-S 

direction each have a minimum length of about 8 ft or more (small 

segments not included). Assuming a design unit shear value of 80 plf / 2 = 

40 plf (safety factor of 2), the design lateral resistance may be at least 40 ft 

x 40 plf = 1,600 lb. Although that is not a large amount, it should factor 

into the design consideration, particularly when a lateral design solution is 

considered to be marginal, based on an analysis that does not consider 

interior partition walls.

2. Given the lower wind shear load in the E-W direction, the identical 

seismic story shear load in the E-W direction, and the greater available 

length of shear wall in the E-W direction, an adequate amount of lateral 

resistance should be no problem for shear walls in the E-W direction. 

Some of the available E-W shear wall segments may not even be required 

to be designed and detailed as shear wall segments. Also, with hold-down 

brackets at the ends of the N-S walls that are detailed to anchor a common 

corner stud (to which the corner sheathing panels on each wall are 

fastened with the required panel edge fastening), the E-W walls are 

essentially perforated shear wall lines and may be treated as such in 

evaluating the design shear capacity of the E-W wall lines.

3. The distribution of the house shear wall elements seems to be reasonably 

“even” in this example; however, the garage opening wall could be 

considered a problem if sufficient connection of the garage to the house is 

not provided to prevent the garage from rotating separately from the house 

under the N-S wind or seismic load. Thus, the garage walls and garage 

roof diaphragm should be adequately attached to the house so that the 

garage and house act as a structural unit. The process will be detailed in 

the next part of this example.

2. Determine the shear loads on the N-S shear wall lines, using the relative 

stiffness method and an assumed shear wall construction for the given seismic 

design condition only.

Assume that the shear wall construction will be as follows—

• 7/16-inch OSB Structural I wood structural panel sheathing with 8d 

common nails (or 0.131-inch diameter 8d pneumatic nails) spaced at 4 

inches on center on the panel edges and 12 inches in the panel field.

• Douglas-fir wall framing is used, with 2x studs spaced at 16 inches on 

center.

• Walls are designed as perforated shear wall lines, and adequate hold-
Residential Structural Design Guide 6-62

downs and base shear connections are provided.

It will be further assumed that the house and garage are sufficiently tied 

together to act as a structural unit. The designer must remember that the 

relative stiffness design approach is predicated on the assumption that the 

horizontal diaphragm is rigid in comparison to the supporting shear walls so 

that the forces are distributed according to the relative stiffness of the shear 

wall lines. 

As given for the design example, the following design seismic shear loads 

apply to the first story of the example building:

Design N-S Seismic Lateral Load (mapped Ss = 1.5g)

House: 7,493 lb total story shear (tributary weight is 37,464 lb)

Garage: 1,490 lb total story shear (tributary weight is 7,452 lb)

Total: 8,983 lb total story shear (total tributary weight is 44,916 lb)

Locate the center of gravity.

The first step is to determine the center of gravity of the building at the first 

story level; the total seismic story shear load will act through this point. For 

wind design, the process is similar, but the horizontal wind forces on various 

portions of the building (based on vertical projected areas and wind pressures) 

are used to determine the force center for the lateral wind loads (that is, the 

resultant of the garage and house lateral wind loads).

Establishing the origin of an x-y coordinate system at the bottom corner of 

wall line B of the example first floor plan, the center of gravity is determined 

by taking weighted moments about each coordinate axis using the center of 

gravity location for the garage and house portions. Again, the “bump-out” area 

in the living room is considered to have negligible impact on the estimate of 

the center of gravity because most of the building mass originates from the 

second story and roof, which do not have the “bump-out” in the plan.

The center of gravity of the garage has the (x, y) coordinates of (-11 ft, 16 ft). 

The center of gravity of the house has the coordinates (21 ft, 14 ft).

Weighted moments about the y-axis—

Xcg,building = [(Xcg,garage)(garage weight) + (Xcg,house)(house 

weight)]/(total weight)

= [(-11 ft)(7,452 lb) + (21 ft)(37,464 lb)]/(44,916 lb)

= 15.7 ft

Weighted moments about the x-axis—

Ycg,building = [(Ycg,garage)(garage weight) + (Ycg,house)(house weight)]/(total 

weight)

= [(16 ft)(7,452 lb) + (14 ft)(37,464 lb)]/(44,916 lb)

= 14.3 ft

Thus, the center of gravity for the first story is located at the (x,y) coordinates 

of (15.7 ft, 14.3 ft). The approximate location on the floor plan is about 4 

inches north of the center bearing wall line and directly in front of the stairwell 

leading down (that is, about 5 feet to the left of the center of the house).

Locate the center of resistance.
Residential Structural Design Guide 6-63

The center of resistance is somewhat more complicated to determine and 

requires an assumption regarding the shear wall stiffness. Two methods of 

estimating the relative stiffness of segmented shear walls are generally 

recognized. One method bases the segmented shear wall stiffness on its length; 

thus, longer shear walls have greater stiffness (and capacity). This method is 

less appealing when multiple segments are included in one wall line, however, 

and particularly when the segments have varying aspect ratios, especially 

narrow aspect ratios, which affect stiffness disproportionately to the length.

The second method bases the segmented shear wall stiffness on the shear 

capacity of the segment, which is more appealing when various shear wall 

constructions are used with variable unit shear values and when variable aspect 

ratios are used, particularly when the unit shear strength is corrected for 

narrow aspect ratios. The method based on strength also is appropriate to use 

with the perforated shear wall method because the length of a perforated shear 

wall has little to do with its stiffness or strength. Rather, the number of 

openings in the wall (as well as its construction) governs its stiffness and 

capacity. The method used in this example, therefore, will use the capacity of 

the perforated shear wall lines as a measure of relative stiffness. The same 

technique may be used with a segmented shear wall design method by 

determining the shear capacity of each shear wall line (comprising one or more 

shear wall segments), as shown in example 6.1.

First, the strength of each shear wall line in the building must be determined. 

Using the perforated shear wall method and the assumed wall construction 

given at the beginning of step 2, the design shear wall line capacities (see 

below) are determined for each of the exterior shear wall lines in the building. 

The window and door opening sizes are shown on the plan so that the 

perforated shear wall calculations can be done, as demonstrated in example 

6.2. It is assumed that no interior shear wall lines will be used (except at the 

shared wall between the garage and the house) and that the contribution of the 

interior partition walls to the stiffness of the building is negligible. As 

mentioned, this assumption can overlook a significant factor in the lateral 

resistance and stiffness of a typical residential building.

PSW 1: Fpsw1 = 7,812 lb (wall line D)

PSW 2: Fpsw2 = 3,046 lb (wall line E)

PSW 3: Fpsw3 = 14,463 lb (north side wall of house)

PSW 4: Fpsw4 = 9,453 lb (north side of garage)

PSW 5: Fpsw5 = 182 lb (wall line A; garage opening)

PSW 6: Fpsw6 = 9,453 lb (south side wall of garage)

PSW 7: Fpsw7 = 9,687 lb (wall line B)

PSW 8: Fpsw8 = 11,015 lb (south side wall of house at front)

The center of stiffness on the y-coordinate is now determined as follows, using 

the above PSW design shear capacities for wall lines oriented in the E-W 

direction—

Ycs = [(Fpsw3)(Ypsw3) + (Fpsw4)(Ypsw4) + (Fpsw6)(Ypsw6) + (Fpsw8)(Ypsw8)]/(Fpsw,EW)

= [(14,463 lb)(28 ft)+(9,453 lb)(26 ft)+(9,453 lb)(6 ft)+(11,015 lb)(0 

ft)]/(44,384 lb)

= 15.9 ft

The center of stiffness on the x-coordinate is determined similarly, considering 

the wall lines oriented in the N-S direction—
Residential Structural Design Guide 6-64

Xcs = [(Fpsw1)(Xpsw1) + (Fpsw2)(Xpsw2) + (Fpsw5)(Xpsw5) + (Fpsw7)(Xpsw7)]/(Fpsw,NS)

= [(7,812 lb)(42 ft)+(3,046 lb)(48 ft)+(182 lb)(-22 ft)+(9,687 lb)(0 

ft)]/(20,727 lb)

= 22.7 ft

The coordinates of the center of stiffness, therefore, are 22.7 ft, 15.9 ft. Thus, 

the center of stiffness is located to the right of the center of gravity (force 

center for the seismic load) by 22.7 ft – 15.7 ft = 7 ft. This offset between the 

center of gravity and the center of resistance will create a torsional response in 

the N-S seismic load direction under consideration. For E-W seismic load 

direction, the offset (in the y-coordinate direction) is only 15.9 ft – 14.3 ft = 

1.6 ft, which is practically negligible from the standpoint of torsional response. 

The designer should remember that, in both loading directions, the influence of 

interior partitions on the center of stiffness (and thus the influence on torsional 

response) is not considered. To conservatively account for this condition and 

for possible error in locating the actual center of gravity of the building (that is, 

accidental torsion), codes usually require that the distance between the center 

of gravity and the center of stiffness be considered as a minimum of 5 percent 

of the building dimension perpendicular to the direction of seismic force under 

consideration. This condition is essentially met in this example because the 

offset dimension for the N-S load direction is 7 feet, which is 10 percent of the 

E-W plan dimension of the house and attached garage.

Distribute the direct shear forces to N-S walls.

The direct shear force is distributed to the N-S walls based on their relative 

stiffness, without regard to the location of the center of stiffness (resistance) 

and the center of gravity (seismic force center), or the torsional load 

distribution that occurs when they are offset from each other. The torsional 

load distribution is superimposed on the direct shear forces on the shear wall 

lines in the next step of the process.

The direct seismic shear force of 8,983 lb is distributed as shown herein, based 

on the relative stiffness of the perforated shear wall lines in the N-S direction. 

As before, the relative stiffness is based on the design shear capacity of each 

perforated shear wall line relative to that of the total design capacity of the N-S 

shear wall lines.

Direct shear on PSW1, PSW2, PSW5, and PSW7 is determined as follows—

(total seismic shear load on story)[(Fpsw1)/(Fpsw,N-S)] = (8,983 lb)[(7,812 

lb)/(20,727 lb)]

= (8,983 lb)[0.377]

= 3,387 lb

(total seismic shear load on story)[(Fpsw2)/(Fpsw,N-S)] = (8,983 lb)[(3,046 

lb)/(20,727 lb)]

= (8,983 lb)[0.147]

= 1,321 lb

(total seismic shear load on story)[(Fpsw5)/(Fpsw,N-S)] = (8,983 lb)[(182 

lb)/(20,727 lb)]

= (8,983 lb)[0.009]

= 81 lb
Residential Structural Design Guide 6-65

(total seismic shear load on story)[(Fpsw7)/(Fpsw,N-S)] = (8,983 lb)[(9,687 

lb)/(20,727 lb)]

= (8,983 lb)[0.467]

= 4,195 lb

Distribute the torsion load.

The torsional moment is created by the offset of the center of gravity (seismic 

force center) from the center of stiffness or resistance (also called the center of 

rigidity). For the N-S load direction, the torsional moment is equal to the total 

seismic shear load on the story, multiplied by the x-coordinate offset of the 

center of gravity and the center of stiffness (that is, 8,983 lb x 7 ft = 62,881 ftlb). The sharing of this torsional moment on all of the shear wall lines is based 

on the torsional moment of resistance of each wall line. The torsional moment 

of resistance is determined by the design shear capacity of each wall line (used 

as the measure of relative stiffness) multiplied by the square of its distance 

from the center of stiffness. The amount of the torsional shear load (torsional 

moment) distributed to each wall line is then determined by each wall’s 

torsional moment of resistance in proportion to the total torsional moment of 

resistance of all shear wall lines combined. The torsional moment of resistance 

of each shear wall line and the total for all shear wall lines (torsional moment 

of inertia) is determined as shown below.

Wall Line Fpsw Distance from Center 

of Resistance

Fpsw(d)

2

PSW1 7,812 lb 19.3 ft 2.91 x 106 lb-ft2

PSW2 3,046 lb 25.3 ft 1.95 x 106 lb-ft2

PSW3 14,463 lb 12.1 ft 2.12 x 106 lb-ft2

PSW4 9,453 lb 10.1 ft 9.64 x 105 lb-ft2

PSW5 182 lb 44.7 ft 3.64 x 105 lb-ft2

PSW6 9,453 lb 9.9 ft 9.26 x 105 lb-ft2

PSW7 9,687 lb 22.7 ft 4.99 x 106 lb-ft2

PSW8 11,015 lb 15.9 ft 2.78 x 106 lb-ft2

Total torsional moment of inertia (J) 1.70 x 107 lb-ft2

The torsional shear load on each wall is determined using the following basic 

equation for torsion—

where,

VWALL = the torsional shear load on the wall line (lb)

MT = the torsional moment* (lb-ft)

d = the distance of the wall from the center of stiffness (ft)

FWALL = the design shear capacity of the segmented or perforated shear

wall line (lb)

J = the torsional moment of inertia for the story (lb-ft2

)

*The torsional moment is determined by multiplying the design shear load on 

the story by the offset of the center of stiffness relative to the center of gravity, 

perpendicular to the load direction under consideration. For wind design, the 

center of the vertical projected area of the building is used in lieu of the center 

of gravity.

J

M d(F ) V T WALL

WALL =
Residential Structural Design Guide 6-66

Now the torsional loads may be determined as follows for the N-S and E-W 

wall lines. For PSW1 and PSW2, the torsion load is in the reverse direction of 

the direct shear load on these walls. This behavior is the result of the center of 

shear resistance being offset from the force center, which causes rotation 

around the center of stiffness (center of shear resistance and center of stiffness

may be used interchangeably because the shear resistance is assumed to 

represent stiffness). If the estimated offset of the center of gravity and the 

center of stiffness is reasonably correct, then the torsional response will tend to 

reduce the shear load on PSW1 and PSW2; however, codes generally do not 

allow the direct shear load on a wall line to be reduced because of torsion; only 

increases should be considered.

The following values for use in the torsion equation apply to this example—

MT = (8,983 lb)(7 ft) = 62,881 ft-lb

J = 1.70 x 107 lb-ft2

The torsional loads on PSW5 and PSW7 are determined as follows—

Vpsw5 = (62,881 ft-lb)(44.7 ft)(182 lb) / (1.70 x 107 lb-ft2

)

= 30 lb

Vpsw7 = (62,881 ft-lb)(22.7 ft)(9,687 lb) / (1.70 x 107 lb-ft2

)

= 813 lb

These torsional shear loads are added to the direct shear loads for the N-S 

walls, and the total design shear load on each wall line may be compared to its 

design shear capacity as follows—

N-S

Wall Lines

Wall 

Design 

Capacity, 

Fpsw (lb)

Direct 

Shear 

Load (lb)

Torsional 

Shear Load 

(lb)

Total 

Design 

Shear 

Load (lb)

Percent 

of 

Design 

Capacity 

Used

PSW1 7,812 3,387 n/a* 3,387 43% 

(OK)

PSW2 3,046 1,321 n/a* 1,321 43% 

(OK)

PSW5 182 81 30 111 61% 

(OK)

PSW7 9,687 4,195 813 5,008 52% 

(OK)

*The torsional shear load is actually in the reverse direction of the direct shear load for these walls, 

but it is not subtracted, as required by code.

Although all of the N-S shear wall lines have sufficient design capacity, the 

wall lines on the left side (west) of the building are noticeably “working 

harder,” and the walls on the right side (east) of the building are substantially 

overdesigned. The wall construction could be changed to allow greater 

sheathing nail spacing on walls PSW1 and PSW2. Also, the assumption of a 

rigid diaphragm over the entire expanse of the story is questionable, even if the 

garage is “rigidly” attached to the house with adequate connections. The loads 

on walls PSW5 and PSW7 will likely be higher than predicted using the 

relative stiffness method. Certainly, reducing the shear wall construction based 

on the above analysis is not recommended prior to “viewing” the design from 

the perspective of the tributary area approach. Similarly, the garage opening 

wall (PSW5) should not be assumed to be adequate based only on the 

preceding analysis in view of the inherent assumptions of the relative stiffness 
Residential Structural Design Guide 6-67

method regarding the horizontal distribution of shear forces. For more compact 

buildings with continuous horizontal diaphragms extending over the entire area 

of each story, the method is less presumptive; however, this qualitative 

observation is true of all of the force distribution methods demonstrated in this 

design example.

Conclusion This seemingly simple design example has demonstrated the many decisions, 

variables, and assumptions to consider in designing the lateral resistance of a 

light-frame home. An experienced designer may come to favor certain options 

or standardized solutions and develop them for repeated use in similar 

conditions. Also, an experienced designer may be able to effectively design 

using simplified analytical methods (that is, the total shear approach shown in 

step 1), supplemented with judgment and detailed evaluations of certain 

constructions or unique details, as appropriate.

This example indicates that a 7/16-inch thick Structural I wood structural panel 

sheathing can be used for all shear wall construction to resist the required wind 

shear loading. A constant sheathing panel edge nail spacing also is possible by 

spacing 3 inches on center if the perforated shear wall method is used and 2 

inches on center if the segmented shear wall method is used (based on the 

worst case condition of wall line B). The wall sheathing nails specified were 

8d pneumatic nails with a 0.113-inch diameter. In general, this wall 

construction will be conservative for most wall lines on the first story of the 

example house. If the seismic shear load were the only factor (that is, the wind 

load condition was substantially less than assumed), the wall construction 

could be simplified even more such that a perforated shear wall design 

approach with a single sheathing fastening requirement may be suitable for all 

shear wall lines except for the garage opening wall.

Finally, designers should avoid numerous variations in construction detailing 

in a single project because it may lead to confusion and error in the field. 

Fewer changes in assembly requirements, fewer parts, and fewer special details 

should all be as important to the design objectives as meeting the required 

design loads. When the final calculation is done (regardless of the complexity 

or simplicity of the analytic approach chosen and the associated uncertainties 

or assumptions), the designer should exercise judgment in making reasonable 

final adjustments to the design to achieve a practical, well-balanced design. As 

a critical final consideration, the designer should be confident that the various 

parts of the structural system are adequately “tied together” to act as a 

structural unit in resisting the lateral loads. This consideration is as much a 

matter of judgment as it is a matter of analysis.
Residential Structural Design Guide 6-68

6.7 References

Ambrose, J., and D. Vergun. 1987. Design for Lateral Forces. New York: John 

Wiley & Sons.

American Iron and Steel Institute (AISI). 1996. Shear Wall Values for Light 

Weight Steel Framing. Publication RP96-19601. Washington, DC: AISI.

American Society of Testing and Materials (ASTM). 2010. Standard Methods of 

Conducting Strength Tests of Panels for Building Construction, ASTM 

E72-13a. West Conshohocken, PA: ASTM.

———. 2012. Standard Practice for Static Load Test for Shear Resistance of 

Framed Walls for Buildings, ASTM E-564, West Conshohocken, PA. 

ASTM.

American Wood Council (AWC). 2008. National Design Specification for Wood

Construction—Wind and Seismic Supplement. American Wood Council, 

Washington, DC.

———. 2012. National Design Specification for Wood Construction, 

ANSI/AF&PA NDS-2012. Washington, DC: AWC.

APA–The Engineered Wood Association. 1998. Nail Penetration and Framing 

Specific Gravity Factors for Shear Walls and Diaphragms, APA Report 

T98-22. Tacoma, WA: APA.

———. 2007. Diaphragms and Shear Walls. Tacoma, WA: APA.

———. 2008. Brace Walls with Wood. Tacoma, WA: APA.

Boughton, G.N., and G.F. Reardon. 1984. “Simulated Cyclone Wind Tests on a 

Timber Framed House.” In Proceedings of the Pacific Timber Engineering 

Conference, Vol. 2. Auckland, New Zealand: New Zealand Forest 

Research Institute: 527–534.

Christovasilis, I.P., A. Filiatrault, and A. Wanitkorkal. 2007. “Seismic Test of a 

Full-Scale Two-Story Light-Frame Wood Building: The NEESWood 

Benchmark Test.” NEESWood report NW-01, MCEER, Buffalo, NY.

Crandell, Jay H., and Zeno Martin. 2009. “The Story Behind the 2009 IRC Wall 

Bracing Provisions,” Wood Design Focus 19 (1).

Diekmann, Edward R. 1986. “Design of Wood Diaphragms.” In Wood: 

Engineering Design Concepts: A Compilation of Educational Modules for 

Engineers and Architects at the Fourth Clark C. Heritage Memorial 
Residential Structural Design Guide 6-69

Series on Wood, Madison, WI, August 1982. Clark C. Heritage Memorial 

Series on Wood, Vol. 4. University Park, PA: Materials Education 

Council, Materials Research Laboratory, Pennsylvania State University:

479–547.

Dolan, J.D., and C.P. Heine. 1997a. Monotonic Tests of Wood-Frame Shear Walls 

with Various Openings and Base Restraint Configurations, Timber 

Engineering Report No. TE-1997-001. Blacksburg, VA: Virginia 

Polytechnic Institute and State University.

———. 1997b. Sequential Phased Displacement Cyclic Tests of Wood-Frame 

Shear Walls with Various Openings and Base Restraint Configurations, 

Timber Engineering Report No. TE-1997-002. Blacksburg, VA: Virginia 

Polytechnic Institute and State University.

———. 1997c. Sequential Phased Displacement Tests of Wood-Framed Shear 

Walls with Corners, Timber Engineering Report No. TE-1997-003. 

Blacksburg, VA: Virginia Polytechnic Institute and State University.

Dolan, J.D., and A.C. Johnson. 1996a. Cyclic Tests of Long Shear Walls with

Openings, Report TE-1996-002. Blacksburg, VA: Virginia Polytechnic 

Institute and State University.

———. 1996b. Monotonic Tests of Long Shear Walls with Openings, Report TE1996-001. Blacksburg, VA: Virginia Polytechnic Institute and State 

University.

Federal Emergency Management Agency. 2003. NEHRP Recommended 

Provisions and Commentary for Seismic Regulations for New Buildings 

and Other Structures. FEMA 450. Washington, DC: FEMA.

———. 2012. Seismic Evaluation and Retrofit of Multi-Unit Wood-Frame 

Buildings with Weak First Stories, FEMA P-807. Washington, DC: 

FEMA.

Filiatrault, A., Christovasilis, I.P., Wanitkorkal, A., and van de Lindt, J.W. 2010. 

“Experimental Seismic Response of a Full-Scale Light-Frame Wood 

Building.” Journal of Structural Engineering, 136(3), 246-254.

Filiatrault, A., D. Fischer, B. Folz, and C.-M. Uang. 2002. “Seismic Testing of 

Two-Story Woodframe House: Influence of Wall Finish Materials,” 

Journal of Structural Engineering 128 (1): 1337–1345.

Filiatrault, A., and R.O. Foschi. 1991. “Static and Dynamic Tests of Timber Shear 

Walls Fastened with Nails and Wood Adhesive,” Canadian Journal of 

Civil Engineering 18 (5): 749–755.
Residential Structural Design Guide 6-70

Griffiths, D.K., and H.G. Wickens. 1996. “Timber Frame Walls: Design for 

Racking Resistance.” In Proceedings of the International Wood 

Engineering Conference, Vol. 2, edited by Vijaya K. A. Gopu. Madison, 

WI: Omnipress: 37-44.

U.S. Department of Housing and Urban Development (HUD). 1999. Evaluation 

of Housing Performance and Seismic Design Parameters in the 

Northridge Earthquake. Report prepared for the National Association of 

Home Builders and HUD. Washington, DC: NAHB Research Center, Inc.

International Code Council, Inc. (ICC). 2012a. International Building Code. 

Washington, DC: ICC.

———. 2012b. International Residential Code. Washington, DC: ICC.

International Code Council Evaluation Services (ICC-ES). 2013. Power-Driven 

Staples and Nails, ESR-1539. Whittier, CA, ICC-ES.

International Council of Building Officials (ICBO). 1969. Application for 

Research Committee Recommendation on Alternate Materials or Methods 

of Construction, K-Lath Corporations, Whittier, CA.

———. 1988. Uniform Building Code, Whittier, CA.

———. 1997. Uniform Building Code, Whittier, CA.

Karacabeyli, E., and A. Ceccotti. 1998. Nailed Wood-Frame Shear Walls for 

Seismic Loads: Test Results and Design Considerations, Paper T207-6, 

Proceedings of the Structural Engineers’ World Congress. Reston, VA: 

American Society of Civil Engineers.

Kasal, B., R.J. Leichti, and R.Y. Itani. 1994. “Non-Linear Finite Element Model 

of Complete Light Frame Wood Structures,” Journal of Structural 

Engineering 120 (1): 100–119.

Krawinkler, H., Parisi, F., Ibarra, L., Ayoub, A., and Medina, R., 2000. 

“Development of a Testing Protocol for Woodframe Structures” CUREE 

Publication No. W-02

Li, M., F. Lam, B.J. Yeh, T. Skaggs, D. Rammer and J. Wacker. 2012. “Modeling 

Force Transfer Around Openings in Wood-Frame Shear Walls,” ASCE 

Journal of Structural Engineering 138 (12): 1419-1426.
Residential Structural Design Guide 6-71

Mosalam, K.M., and S.A. Mahin. 2007. “Seismic Evaluation and Retrofit of 

Asymmetric Multi-Story Wood-Frame Building,” Journal of Earthquake 

Engineering 11 (6): 968–986.

National Association of Home Builders (NAHB). 1997. Monotonic Tests of ColdFormed Steel Shear Walls with Openings. Report prepared for the U.S. 

Department of Housing and Urban Development, American Iron and Steel 

Institute, and NAHB. Upper Marlboro, MD: NAHB Research Center, Inc.

———. 1998. The Performance of Perforated Shear Walls with Narrow Wall 

Segments, Reduced Base Restraint, and Alternative Framing Methods. 

Report prepared for the U.S. Department of Housing and Urban 

Development. Upper Marlboro, MD: NAHB Research Center, Inc.

———. 1999. Perforated Shear Walls with Conventional and Innovative Base 

Restraint Connections. Report prepared for the U.S. Department of 

Housing and Urban Development. Upper Marlboro, MD: NAHB Research 

Center, Inc.

———. 2009. Evaluation of Full-Scale House Testing Under Lateral Loading, 

Report #5822-03_01162009. Upper Marlboro, MD: NAHB Research 

Center, Inc.

National Bureau of Standards (NBS). 1948. Strength of Houses: Application of 

Engineering Principles to Structural Design, Building Materials and 

Structures Report 109. Washington, DC: U.S. Department of Commerce 

and U.S. Government Printing Office.

National Institute of Standards and Technology (NIST). 1998. A Summary of the 

Structural Performance of Single-Family, Wood-Frame Housing, NISTIR 

6224. Gaithersburg, MD: NIST.

Ni, C., E. Karacabcyli, and A. Ceccotti, D. 1999. Design of Shear Walls with 

Openings under Lateral and Vertical Loads.” in Proceedings of the Pacific 

Timber Engineering, Rotorua, New Zealand Forest Research Institute: 

144-218. 

Patton-Mallory, M., R.M. Gutkowski, and L.A. Soltis. 1984. Racking 

Performance of Light Frame Walls Sheathed on Two Sides, Research 

Paper FPL, Vol. 448. Madison, WI: U.S. Department of Agriculture, 

Forest Service, and Forest Products Laboratory.

Pei, S., J.W. van de Lindt, S.E. Pryor, H. Shimizu, H. Isoda, and D. Rammer. 

2010. Seismic Testing of a Full-Scale Mid-Rise Building: The NEESWood 

Capstone Test, NEESWood Project Report NW-04. http://www.nees.org.
Residential Structural Design Guide 6-72

Reardon, G.F. 1988. “Effects of Non-Structural Cladding on Timber Framed 

House Construction.” In Proceedings of the International Conference on 

Timber Engineering, Madison, WI, Forest Products Research Society: 

276-281.

Reardon, G.F. 1989. “Effect of Claddings on the Response of Houses to Wind 

Forces.” In Proceedings of the Second Pacific Timber Engineering 

Conference, Vol. 2. University of Auckland: Auckland, New Zealand: 

101–105.

Reardon, G., and D. Henderson. 1996. “Simulated Wind Loading of a Two-Story 

Test House.” In Proceedings of the International Wood Engineering 

Conference, Vol. 4. Madison, WI: Omnipress: 313-319.

Sugiyama and Matsumoto. 1994. “Empirical Equations for the Estimation of 

Racking Strength of a Plywood-Sheathed Shear Wall with Openings.” In 

Summaries of Technical Papers of Annual Meeting, Architectural Institute 

of Japan.

Testing Engineers, Inc. 1970. Strength Tests of Plastered Building Panels for 

Keystone Steel and Wire Division. Testing Engineers, Inc., Oakland, CA.

———. 1971. Strength Tests of Plastered Building Panels using ASTM E72-61.” 

Testing Engineers, Inc., Oakland, CA.

Thurston, S.J. 1994. Report on Racking Resistance of Long Sheathed Timber 

Framed Walls with Openings. Study Report 54 of Building Research 

Association of New Zealand (BRANZ) Study Report. BRANZ, Judgeford, 

Wellington, New Zealand.

Tuomi, R.L. and W.J. McCutcheon. 1974. Testing of a Full-Scale House Under 

Simulated Snow Loads and Wind Loads, U.S. Department of Agriculture, 

Forest Products Laboratory, Madison, WI.

van de Lindt, J., S. Pei, S.E. Pryor, H. Shimizu, and H. Isoda. 2010. 

“Experimental Seismic Response of a Full-Scale Six-Story Light-Frame 

Wood Building,” ASCE Journal of Structural Engineering 136 (10): 

1262–1272.

van de Lindt, J.W. and R. Gupta. 2006. “Damage and Damage Prediction for 

Wood Shearwalls Subjected to Simulated Earthquake Loads.” ASCE 

Journal of Performance of Constructed Facilities 20 (2), 176-184. 

van de Lindt, J.W., D.V. Rosowsky, W. Pang, and S. Pei. 2013. “PerformanceBased Seismic Design of Midrise Woodframe Buildings,” ASCE Journal 

of Structural Engineering 139 (8): 1294–1302.
Residential Structural Design Guide 6-73

White, M.W., and Dolan, J.D. 1993. “Effect of Adhesives on Shear Wall 

Performance.” In Systems Approach to Wood Structures. Proceedings. 

Madison, WI: Forest Products Society.

Wolfe, R.W. 1983. Contribution of Gypsum Wallboard to Racking Resistance of 

Light Frame Walls, Research Paper FPL 439. Madison, WI: U.S. 

Department of Agriculture, Forest Service, and Forest Products 

Laboratory. 

Yokel, F.Y., G. Hsi, and F. Somes. 1973. Full-Scale Test on a Two-Story House 

Subjected to Lateral Load. Washington, DC: U.S. Department of 

Commerce, National Bureau of Standards.
Residential Structural Design Guide 7-1

CHAPTER 7

Connections

7.1 General

The objectives of connection design are—

• To transfer loads resisted by structural members and systems to other 

parts of the structure to form a “continuous load path.”

• To secure nonstructural components and equipment to the building.

• To fasten members in place during construction to resist temporary 

loads during installation (for example, finishes and sheathing).

Adequate connection of the framing members and structural systems 

(covered in chapters 4, 5, and 6) is a critical design and construction 

consideration. Regardless of the type of structure or type of material, structures 

are only as strong as their connections, and structural systems can behave as a unit 

only with proper interconnection of the components and assemblies; therefore, 

this chapter is dedicated to connections. A connection transfers loads from one 

framing member to another (for example, a stud to a top or bottom plate) or from 

one assembly to another (for example, a roof to a wall, a wall to a floor, and a 

floor to a foundation). Connections generally consist of two or more framing 

members and a mechanical connection device, such as a fastener or specialty 

connection hardware. Adhesives also are used to supplement mechanical 

attachment of wall finishes or floor sheathing to wood.

This chapter focuses on conventional wood connections that typically use 

nails, bolts, and some specialty hardware. The procedures for designing 

connections are based on the National Design Specification for Wood 

Construction (NDS) (AWC, 2012). The chapter also addresses relevant concrete 

and masonry connections, in accordance with the applicable provisions of 

Building Code Requirements for Structural Concrete (ACI-318) (ACI, 2011) and 
Residential Structural Design Guide 7-2

Building Code Requirements for Masonry Structures (TMS 402) (ACI, 2008). 

When referring to the NDS, ACI-318, or TMS 402, the chapter identifies 

particular sections as NDS•12.1, ACI-318•22.5, or TMS 402.

For most connections in typical residential construction, the connection 

design may be based on prescriptive tables found in the applicable residential 

building code (ICC, 2012) or in other documents referenced in the International 

Residential Code (ICC, 2012), such as the Wood Frame Construction Manual

(WFCM).

The NDS recognizes in NDS•7.1.1.4 that “extensive experience” 

constitutes a reasonable basis for design. The designer may use prescriptive 

nailing schedules for some connections, but he or she must verify that the 

connection complies with local requirements, practice, and the design conditions 

that are assumed for use with the prescriptive schedules. A connection design 

based on the NDS or other sources, such as the WFCM, may be necessary for 

special conditions, such as in high-hazard seismic or wind areas and when unique 

structural details or materials are used. Also, residential framing may incorporate 

large open spaces, vaulted ceilings, steep pitched roofs (greater than a 45-degree 

slope), and other elements that may fall outside the parameters of prescriptive 

designs; connection designs may therefore be required for many of these 

conditions. Some state and local building codes may require nailing patterns and 

connection details that are more restrictive than those set out in prescriptive 

nailing schedules because of local experience with building performance during 

natural hazard events.

In addition to the conventional fasteners mentioned, many specialty 

connectors and fasteners are available on today’s market. The reader is 

encouraged to gather, study, and scrutinize manufacturer literature regarding 

specialty fasteners, connectors, and tools that meet a wide range of connection 

needs.

7.2 Types of Mechanical Fasteners

Mechanical fasteners that are generally used for wood-framed house 

design and construction include the following.

• Nails and spikes.

• Bolts.

• Lag bolts (lag screws).

• Specialty connection hardware.

This section presents some basic descriptions and technical information on 

those fasteners. Sections 7.3 and 7.4 provide design values and related guidance. 

Section 7.5 presents design examples for various typical conditions in residential 

wood framing and foundation construction.
Residential Structural Design Guide 7-3

7.2.1 Nails

Several characteristics distinguish one nail from another. Figure 7.1 

depicts key nail features for a few types of nails that are essential to wood-framed 

design and construction. This section discusses some of a nail’s characteristics 

relative to structural design; the reader is referred to Standard Terminology of 

Nails for Use With Wood and Wood-Base Materials (ASTM F547) and Standard 

Specification for Driven Fasteners: Nails, Spikes, and Staples (ASTM F1667) for 

additional information (ASTM, 2011; 2012).

FIGURE 7.1 Elements of a Nail and Nail Types
Residential Structural Design Guide 7-4

The following are the most common nail types used in residential wood construction.

• Common nails are bright, plain-shank nails with a flat head and 

diamond point. The diameter of a common nail is larger than that of 

sinkers and box nails of the same length. Common nails are used 

primarily for rough framing.

• Sinker nails are bright or coated slender nails with a sinker head and 

diamond point. The diameter of the head is smaller than that of a 

common nail with the same designation. Sinker nails are used 

primarily for rough framing and applications where lumber splitting 

may be a concern.

• Box nails are bright, coated, or galvanized nails with a flat head and 

diamond point. They are made of lighter gauge wire than common 

nails and sinkers and are commonly used for toenailing and many 

other light framing connections where splitting of lumber is a concern.

• Cooler nails are similar to the nails already described but with slightly 

thinner shanks. They are commonly supplied with ring shanks (that is, 

annular threads) for use as a drywall nail.

• Power-driven nails (and staples) are produced by a variety of 

manufacturers for several types of power-driven fasteners. Pneumaticdriven nails and staples are the most popular power-driven fasteners in 

residential construction. Nails are available in a variety of diameters, 

lengths, and head styles. The shanks are generally cement coated, and 

some nails are available with deformed shanks for added capacity. 

Staples also are available in a variety of wire diameters, crown widths, 

and leg lengths. Refer to ESR-1539 for additional information and 

design data (ICC-ES, 2013).

• Ring-shank nails have rings cut into the shank of the nail and are being 

increasingly used in attaching roof or wall sheathing to framing 

members. The rings significantly increase the withdrawal resistance of 

the nail, and the capacity of ring shank nails could be nearly 50 percent 

greater than plain shank nails, depending on nail diameter and the 

wood species into which the nail is driven (AWC, 2012).

Nail lengths and weights are denoted by the penny weight, which is 

indicated by d. Given the standardization of common nails, sinkers, and cooler 

nails, the penny weight also denotes a nail’s head and shank diameter. For other 

nail types, sizes are based on the nail’s length and diameter. Table 7.1 arrays 

dimensions for the nails discussed herein. Nail length and diameter are key factors 

in determining the strength of nailed connections in wood framing. The steel yield 
Residential Structural Design Guide 7-5

strength of the nail may also be important for certain shear connections, yet such 

information is rarely available for a “standard” lot of nails.

TABLE 7.1 Nail Types, Sizes, and Dimensionsa

Type of Nail Nominal Size

(penny weight, d)

Length

(inches)

Diameter

(inches)

Common

6d 2 0.113

8d 2 1/2 0.131

10d 3 0.148

12d 3 1/4 0.148

16d 3 1/2 0.162

20d 4 0.192

Box

6d 2 0.099

8d 2 1/2 0.113

10d 3 0.128

12d 3 1/4 0.128

16d 3 1/2 0.135

Sinker

6d 1 7/8 0.092

8d 2 3/8 0.113

10d 2 7/8 0.120

12d 3 1/8 0.135

16d 3 1/4 0.148

Pneumaticb

6d 1 7/8 to 2 0.092 to 0.113

8d 2 3/8 to 2 1/2 0.092 to 0.131

10d 3 0.120 to 0.148

12d 3 1/4 0.120 to 0.131

16d 3 1/2 0.131 to 0.162

20d 4 0.131

Cooler

4d 1 3/8 0.067

5d 1 5/8 0.080

6d 1 7/8 0.092

Notes:

a Based on ASTM F1667 (ASTM, 2011).

b Based on a survey of pneumatic fastener manufacturer data and ESR-1539 (ICC, 2013).

Although many types of nail heads are available, three types are most 

commonly used in residential wood framing.

• The flat nail head is the most common head. It is flat and circular, and 

its top and bearing surfaces are parallel but with slightly rounded 

edges.

• The sinker nail head is slightly smaller in diameter than the flat nail 

head. It also has a flat top surface; however, the bearing surface of the 

nail head is angled, allowing the head to be slightly countersunk.
Residential Structural Design Guide 7-6

• Pneumatic heads are available in the above types; however, other head 

types, such as half-round or D-shaped heads, also are common.

The shank, as illustrated in figure 7.1, is the main body of a nail. It extends 

from the head of the nail to the point. It may be plain or deformed. A plain shank 

is considered a “smooth” shank, but it may have “grip marks” from the 

manufacturing process. A deformed shank is most often either threaded or fluted 

to provide additional withdrawal or pullout resistance. Threads are annular (that 

is, ring shank), helical, or longitudinal deformations rolled onto the shank, 

creating ridges and depressions. Flutes are helical or vertical deformations rolled 

onto the shank. Threaded nails most often are used to connect wood to wood, 

whereas fluted nails are used to connect wood to concrete (for example, sill plate 

to concrete slab or furring strip to concrete or masonry). Shank diameter and 

surface condition both affect a nail’s capacity.

The nail tip, as illustrated in figure 7.1, is the end of the shank—usually 

tapered—that is formed during manufacturing to expedite nail driving into a given 

material. Among the many types of nail points, the diamond point is most 

commonly used in residential wood construction. The diamond point is a 

symmetrical point with four approximately equal beveled sides that form a 

pyramid shape. A cut point used for concrete cut nails describes a blunt point. The 

point type can affect nail drivability, lumber splitting, and strength characteristics.

The material used to manufacture nails may be steel, stainless steel, heattreated steel, aluminum, or copper, although the most commonly used materials 

are steel, stainless steel, and heat-treated steel. Steel nails typically are formed 

from basic steel wire. Stainless steel nails often are recommended in exposed 

construction near the coast or for certain applications, such as cedar siding, to 

prevent staining. Stainless steel nails also are recommended for permanent wood 

foundations. Heat-treated steel includes annealed, case-hardened, or hardened 

nails that can be driven into particularly hard materials, such as extremely dense 

wood or concrete.

Various nail coatings provide corrosion resistance, increased pullout 

resistance, or ease of driving. Following is a description of some of the more 

common coatings in residential wood construction.

• Bright. Uncoated and clean nail surface.

• Cement-coated. Coated with a heat-sensitive cement that prevents 

corrosion during storage and improves withdrawal strength, depending 

on the moisture and density of the lumber and other factors.

• Galvanized. Coated with zinc by barrel-tumbling, dipping, 

electroplating, flaking, or hot-dipping to provide a corrosion-resistant 

coating during storage and after installation for either performance or 

appearance. The coating thickness increases the diameter of the nail 

and improves withdrawal and shear strength.
Residential Structural Design Guide 7-7

7.2.2 Bolts

Bolts often are used for “heavy” connections and to secure wood to other 

materials, such as steel or concrete. In many construction applications, however, 

special power-driven fasteners are used in place of bolts. Refer to figure 7.2 for an 

illustration of some typical bolt types and connections for residential use.

FIGURE 7.2 Bolt and Connection Types
Residential Structural Design Guide 7-8

In residential wood construction, bolted connections typically are limited 

to wood-to-concrete or wood-to-masonry used for foundation connections, unless 

a home is constructed in a high-hazard wind or seismic area and hold-down 

brackets are required to transfer shear wall overturning forces (see chapter 6). 

Foundation bolts, typically embedded in concrete or grouted masonry, are 

commonly referred to as anchor bolts, J-bolts, headed anchor bolts, or mud-sill 

anchors. Another type of bolt sometimes used in residential construction is the 

structural bolt, which connects wood to steel or wood to wood. Low-strength 

ASTM A307 bolts commonly are used in residential construction, whereas highstrength ASTM A325 bolts are more common in commercial applications. Bolt 

diameters in residential construction generally range from 1/4 to 3/4 inch, 

although 1/2- to 5/8-inch-diameter bolts are most common, particularly for 

connecting a 2x wood sill to grouted masonry or concrete.

Bolts, unlike nails, are installed in predrilled holes. If holes are too small, 

the possibility of splitting the wood member increases during installation of the 

bolt. If bored too large, the bolt holes encourage nonuniform dowel (bolt) bearing 

stresses and slippage of the joint when loaded. NDS•8.1 specifies that bolt holes 

should range from 1/32 to 1/16 inch larger than the bolt diameter to prevent 

splitting and to ensure reasonably uniform dowel bearing stresses.

7.2.3 Specialty Connection Hardware

Many manufacturers fabricate specialty connection hardware. The load 

capacity of a specialty connector usually is obtained through testing to determine 

the required structural design values. The manufacturer’s product catalog 

typically provides the required values; thus, the designer can select a standard 

connector based on the design load determined for a particular joint or connection 

(see chapter 3). The designer should carefully consider the type of fastener to be 

used with the connector, however; sometimes a manufacturer requires or offers 

proprietary nails, screws, or other devices. The designer should also verify the 

safety factor and strength adjustments used by the manufacturer, including the 

basis of the design value. In some cases, as with nailed and bolted connections in 

the NDS, the basis is a serviceability limit state (that is, slip or deformation) and 

not ultimate capacity.

Following is discussion of a few examples of specialty connection 

hardware, which are illustrated in figure 7.3.

• Sill anchors are used in lieu of foundation anchor bolts. Many 

configurations are available in addition to the one shown in figure 7.3.

• Joist hangers are used to attach single or multiple joists to the side of 

girders or header joists.

• Rafter clips and roof tie-downs are straps or brackets that connect roof 

framing members to wall framing to resist roof uplift loads associated 

with high wind conditions.
Residential Structural Design Guide 7-9

• Hold-down brackets are brackets that are bolted, nailed, or screwed to 

wall studs or posts and anchored to the construction below (that is, 

concrete, masonry, or wood) to “hold down” the end of a member or 

assembly (that is, shear wall).

• Strap ties are prepunched straps or coils of strapping that are used for a 

variety of connections to transfer tension loads.

• Splice plates or shear plates are flat plates with prepunched holes for 

fasteners to transfer shear or tension forces across a joint.

• Epoxy-set anchors are anchor bolts that are drilled and installed with 

epoxy adhesives into concrete after the concrete has cured and 

sometimes after the framing is complete so that the required anchor 

location is obvious.
Residential Structural Design Guide 7-10

FIGURE 7.3 Specialty Connector Hardware

7.2.4 Lag Screws

Lag screws are available in the same diameter range as bolts; the principal 

difference between the two types of connectors is that a lag screw has screw 

threads that taper to a point. The threaded portion of the lag screw anchors itself 

in the main member that receives the tip. Lag screws (often called lag bolts) 
Residential Structural Design Guide 7-11

function as bolts in joints where the main member is too thick to be economically 

penetrated by regular bolts. They are also used when one face of the member is 

not accessible for a “through-bolt.” Holes for lag screws must be carefully drilled 

to one diameter and depth for the shank of the lag screw and to a smaller diameter 

for the threaded portion. Self-tapping lag screws, which do not require a pilot hole 

to be drilled, are now on the market. These lag screws also can have higher yield 

strengths to improve shear resistance. Lag screws in residential applications 

generally are small in diameter and may be used to attach garage door tracks to 

wood framing, steel angles to wood framing supporting brick veneer over wall 

openings, various brackets or steel members to wood, and wood ledgers to wall 

framing.

7.3 Wood Connection Design

7.3.1 General

This section covers the NDS design procedures for nails, bolts, and lag 

screws. The procedures are intended for allowable stress design (ASD) such that 

loads should be determined accordingly (see chapter 3). Other types of fastenings 

are addressed by the NDS, but they are rarely used in residential wood 

construction. The applicable sections of the NDS related to connection design as 

covered in this chapter include—

• NDS•7—Mechanical Connections (General Requirements).

• NDS•8—Bolts.

• NDS•9—Lag Screws.

• NDS•12—Nails and Spikes.

Although wood connections generally are responsible for the complex, 

nonlinear behavior of wood structural systems, the design procedures outlined in 

the NDS are straightforward. The NDS connection values generally are 

conservative from a structural safety standpoint. Further, the NDS’s basic or 

tabulated design values are associated with tests of single fasteners in 

standardized conditions. As a result, the NDS provides several adjustments to 

account for various factors that alter the performance of a connection; in 

particular, the performance of wood connections is highly dependent on the 

species (that is, density or specific gravity) of wood. Table 7.2 provides the 

specific gravity values of various wood species typically used in house 

construction.
Residential Structural Design Guide 7-12

TABLE 7.2 Common Framing Lumber Species and Specific Gravity 

Values

Lumber Species Specific Gravity, G

Southern pine (SP) 0.55

Douglas-Fir-Larch (DF-L) 0.50

Hem-Fir (HF) 0.43

Spruce-Pine-Fir (SPF) 0.42

Spruce-Pine-Fir (South) 0.36

The moisture condition of the wood also is critical to long-term 

connection performance, particularly for nails during withdrawal. In some cases, 

the withdrawal value of fasteners installed in moist lumber can decrease by as 

much as 50 percent over time, as the lumber dries to its equilibrium moisture 

content. At the same time, a nail may develop a layer of rust that increases 

withdrawal capacity. In contrast, deformed shank nails tend to hold their 

withdrawal capacity much more reliably under varying moisture and use 

conditions. For this and other reasons, the design nail withdrawal capacities in the 

NDS for smooth shank nails are based on a fairly conservative reduction factor, 

resulting in about one-fifth of the average ultimate tested withdrawal capacity. 

The reduction includes a safety factor as well as a load duration adjustment (that 

is, decreased by a factor of 1.6 to adjust from short-term tests to a normal duration 

load). Design values for nails and bolts in shear are based on a deformation (that 

is, slip) limit state, not their ultimate capacity, resulting in a safety factor that may 

range from 3 to 5, based on ultimate tested capacities. One argument for retaining 

a high safety factor in shear connections is that the joint may creep under longterm load. Although creep is not a concern for many joints, slip of joints in a 

trussed assembly (such as a rafter-ceiling joist roof framing) is critical and, in key 

joints, can result in a magnified deflection of the assembly over time (that is, 

creep).

As the preceding discussion indicates, a number of uncertainties in the 

design of connections can lead to conservative or unconservative designs relative 

to the intent of the NDS and practical experience. The designer is advised to 

follow the NDS procedures carefully but should be prepared to make practical 

adjustments as dictated by sound judgment and experience and allowed in the 

NDS (refer to NDS•7.1.1.4).

Withdrawal design values for nails and lag screws in the NDS are based 

on the fastener being oriented perpendicular to the grain of the wood. Shear 

design values in wood connections also are based on the fastener being oriented 

perpendicular to the grain of wood; however, the lateral (shear) design values are 

dependent on the direction of loading relative to the wood grain direction in each 

of the connected members. Figure 7.4 illustrates various connection types and 

loading conditions.
Residential Structural Design Guide 7-13

FIGURE 7.4 Types of Connections and Loading Conditions

The NDS provides tabulated connection design values that use the 

following symbols for the three basic types of loading.

• W—withdrawal (or tension loading).

• Z⊥—shear perpendicular to wood grain.

• Z||—shear parallel to wood grain.
Residential Structural Design Guide 7-14

In addition to the already tabulated design values for the preceding 

structural resistance properties of connections, the NDS provides calculation 

methods to address conditions that may not be covered by the tables and that give 

more flexibility to the design of connections. The methods are appropriate for use 

in hand calculations or with computer spreadsheets.

For withdrawal, the design equations are relatively simple empirical 

relationships (based on test data) that explain the effect of fastener size 

(diameter), penetration into the wood, and density of the wood. For shear, the 

equations are somewhat more complex because of the multiple failure modes that 

may result from fastener characteristics, wood density, and size of the wood 

members. Six shear-yielding modes (and a design equation for each) address 

various yielding conditions in either the wood members or the fasteners that join 

the members. The critical yield mode is used to determine the design shear value 

for the connection (refer to NDS•appendix I for a description of the yield modes).

The yield equations in the NDS are based on general dowel equations that 

use principles of engineering mechanics to predict the shear capacity of a doweled 

joint. The general dowel equations can be used with joints that have a gap 

between the members, and they can also be used to predict ultimate capacity of a 

joint made of wood, wood and metal, or wood and concrete (AWC, 2014). The 

equations do not account for friction between members, however, or for the 

anchoring or cinching effect of the fastener head as the joint deforms and the 

fastener rotates or develops tensile forces. These effects are important to the 

ultimate capacity of wood connections in shear; therefore, the general dowel 

equations may be considered to be conservative (refer to section 7.3.6). For 

additional guidance and background on the use of the general dowel equations, 

refer to the NDS Commentary and other useful design resources available through 

the American Wood Council (AWC, 2012; Showalter, Line, and Douglas, 1999).

7.3.2 Adjusted Allowable Design Values

Design values for wood connections are subject to adjustments in a 

manner similar to that required for wood members themselves (see section 5.2.4 

of chapter 5). The calculated or tabulated design values for W and Z are 

multiplied by the applicable adjustment factors to determine adjusted allowable 

design values, W' and Z', as shown in the following extract from the NDS, for the 

various connection methods (that is, nails, bolts, and lag screws).

NDS•12.3 & 7.3

 for bolts.

 for lag screws.

 for nails and spikes.

NDS•12.2 & 7.3

 for nails and spikes.

 for lag screws.

= ∆ Z′ ZCDCM CtCgC

Z = ZCDCM CtCgC∆CdCeg ′

Z = ZCDCM CtCdCegCdiCtn ′

W = WCDCM CtCtn ′

W = WCDCM CtCeg ′
Residential Structural Design Guide 7-15

The adjustment factors and their applicability to wood connection design are briefly 

described as follows.

• CD—Load Duration Factor (NDS•2.3.2 and chapter 5, table 5.2)—applies to W and 

Z values for all fasteners, based on design load duration, but shall not exceed 1.6 (that 

is, wind and earthquake load duration factor).

• CM—Wet Service Factor (NDS•7.3.3)—applies to W and Z values for all connections 

based on moisture conditions at the time of fabrication and during service; not 

applicable to residential framing.

• Ct—Temperature Factor (NDS•7.3.4)—applies to the W and Z values for all 

connections exposed to sustained temperatures of greater than 100oF; not typically 

used in residential framing.

• Cg—Group Action Factor (NDS•7.3.6)—applies to Z values of two or more bolts or 

lag screws loaded in single or multiple shear and aligned in the direction of the load 

(that is, rows).

• C∆—Geometry Factor (NDS•8.5.2, 9.4.)—applies to the Z values for bolts and lag 

screws when the end distance or spacing of the bolts is less than assumed in the 

unadjusted design values.

• Cd—Penetration Depth Factor (NDS•9.3.3, 12.3.4)—applies to the Z values of lag 

screws and nails when the penetration into the main member is less than 8D for lag 

screws or 12D for nails (where D = shank diameter); sometimes applicable to 

residential nailed connections.

• Ceg—End Grain Factor (NDS•9.2.2, 9.3.4, 12.3.5)—applies to W and Z values for 

lag screws and to Z values for nails to account for reduced capacity when the fastener 

is inserted into the end grain (Ceg = 0.67).

• Cdi—Diaphragm Factor (NDS•12.3.6)—applies to the Z values of nails only to 

account for system effects from multiple nails used in sheathed diaphragm 

construction (Cdi = 1.1).

• Ctn—Toenail Factor (NDS•12.3.7)—applies to the W and Z values of toenailed 

connections (Ctn = 0.67 for withdrawal and = 0.83 for shear). It does not apply to slant 

nailing in withdrawal or shear; refer to section 7.3.6.

The total allowable design value for a connection (as adjusted by the 

appropriate factors above) must meet or exceed the design load determined for the 

connection (refer to chapter 3 for design loads). The values for W and Z are based 

on single fastener connections. In instances of connections involving multiple 

fasteners, the values for the individual or single fastener can be summed to 

determine the total connection design value only when Cg is applied (to bolts and 
Residential Structural Design Guide 7-16

lag screws only) and fasteners are the same type and similar size. This approach 

may overlook certain system effects that can improve the actual performance of 

the joint in a constructed system or assembly, however (see section 7.3.6). 

Conditions that may decrease estimated performance, such as prying action 

induced by the joint configuration or eccentric loads and other factors, should also 

be considered.

In addition, the NDS does not provide values for nail withdrawal or shear 

when wood structural panel members (that is, plywood or oriented strand board) 

are used as a part of the joint. This type of joint—wood member to structural 

wood panel—occurs frequently in residential construction. Z values can be 

estimated by using the yield equations for nails in NDS 12.3.1 and assuming a 

reasonable specific gravity (density) value for the wood structural panels, such as 

G = 0.5. W values for nails in wood structural panels can be estimated in a similar 

fashion by using the withdrawal equation presented in the next section. The 

tabulated W and Z values in NDS•12 may also be used but with some caution as 

to the selected table parameters.

7.3.3 Nailed Connections

The procedures in NDS•12 provide for the design of nailed connections to 

resist shear and withdrawal loads in wood-to-wood and metal-to-wood 

connections. As mentioned, many specialty “nail-type” fasteners are available for 

wood-to-concrete and even wood-to-steel connections. The designer should 

consult manufacturer data for connection designs that use proprietary fastening 

systems.

The withdrawal strength of a smooth nail (driven into the side grain of 

lumber) is determined in accordance with either the empirical design equation that 

follows or NDS•table 12.2A.

NDS•12.2.1

 unadjusted withdrawal design value (lb) for a smooth shank 

nail.

where,

G = specific gravity of the lumber member receiving the nail tip.

D = the diameter of the nail shank (in).

Lp = the depth of penetration (in) of the nail into the member receiving the nail tip.

The design strength of nails is greater when a nail is driven into the side 

rather than the end grain of a member. Withdrawal information is available for 

nails driven into the side grain; however, the withdrawal capacity of a nail driven 

into the end grain is assumed to be zero because of its unreliability. Furthermore, 

the NDS does not provide a method for determining withdrawal values for 

deformed shank nails. These nails significantly enhance withdrawal capacity and 

are frequently used to attach roof sheathing in high-wind areas. They also are used 

to attach floor sheathing and some siding materials to prevent nail “back-out.” 

The use of deformed shank nails usually is based on experience or preference. 

p

2

5

W = 1380(G) DL
Residential Structural Design Guide 7-17

Withdrawal values for deformed shank nails are provided in ICC-ES Report 1539 

(ICC-ES, 2013) on pneumatically driven fasteners.

The design shear value, Z, for a nail typically is determined using the 

following tables from NDS•12.

• Tables 12.3A and B. Nailed wood-to-wood, single-shear (twomember) connections with the same species of lumber using box or 

common nails, respectively.

• Tables 12.3E and F. Nailed metal plate-to-wood connections using 

box or common nails, respectively.

The yield equations in NDS•12.3 may be used for conditions not 

represented in the design value tables for Z. Regardless of the method used to 

determine the Z value for a single nail, the value must be adjusted as described in 

section 7.3.2. As noted in the NDS, the single nail value is used to determine the 

design value.

It is also worth mentioning that the NDS provides an equation for 

determining allowable design value for shear when a nailed connection is loaded 

in combined withdrawal and shear (see NDS•12.3.8, equation 12.3-6). The 

equation appears to be most applicable to a gable-end truss connection to the roof 

sheathing under conditions of roof sheathing uplift and wall lateral load owing to 

wind. The designer might contemplate other applications but should take care in 

considering the combination of loads that would be necessary to create 

simultaneous uplift and shear worthy of a special calculation.

7.3.4 Bolted Connections

Bolts may be designed in accordance with NDS•8 to resist shear loads in 

wood-to-wood, wood-to-metal, and wood-to-concrete connections. As mentioned, 

many specialty “bolt-type” fasteners can be used to connect wood to other 

materials, particularly concrete and masonry. One common example is an epoxyset anchor. Manufacturer data should be consulted for connection designs that use 

proprietary fastening systems.

The design shear value Z for a bolted connection is typically determined 

by using the following tables from NDS•8.

• Table 8.2A. Bolted wood-to-wood, single-shear (two-member) 

connections with the same species of lumber.

• Table 8.2B. Bolted metal plate-to-wood, single-shear (two-member) 

connections; metal plate thickness of 1/4 inch minimum.

• Table 8.2D. Bolted single-shear wood-to-concrete connections; based 

on minimum 6-inch bolt embedment in minimum fc = 2,000 psi 

concrete.
Residential Structural Design Guide 7-18

The yield equations of NDS•8.2 (single-shear joints) and NDS•8.3 

(double-shear joints) may be used for conditions not represented in the design 

value tables. Regardless of the method used to determine the Z value for a single 

bolted connection, the value must be adjusted as described in section 7.3.2.

It should be noted that the NDS does not provide W values for bolts. The 

tension value of a bolt connection in wood framing is usually limited by the 

bearing capacity of the wood as determined by the surface area of a washer used 

underneath the bolt head or nut. When calculating the bearing capacity of the 

wood based on the tension in a bolted joint, the designer should use the small 

bearing area value Cb to adjust the allowable compressive stress perpendicular to 

grain Fc⊥ (see NDS•2.3.10). It should also be remembered that the allowable 

compressive stress of lumber is based on a deformation limit state, not capacity; 

refer to section 5.2.3 of chapter 5. In addition, the designer should verify the 

tension capacity of the bolt and its connection to other materials (that is, concrete 

or masonry, as covered in section 7.4). The bending capacity of the washer should 

also be considered. For example, a wide but thin washer will not evenly distribute 

the bearing force to the surrounding wood.

The arrangement of bolts and drilling of holes are extremely important to 

the performance of a bolted connection. The designer should carefully follow the 

minimum edge, end, and spacing requirements of NDS•8.5. When necessary, the 

designer should adjust the design value for the bolts in a connection by using the 

geometry factor C and the group action factor Cg discussed in section 7.3.2.

Any possible torsional load on a bolted connection (or any connection for 

that matter) should also be considered in accordance with the NDS. In such 

conditions, the pattern of the fasteners in the connection can become critical to 

performance in resisting both a direct shear load and the loads created by a 

torsional moment on the connection. Fortunately, this condition is not often 

applicable to typical light-frame construction. However, cantilevered members 

that rely on connections to “anchor” the cantilevered member to other members 

will experience this effect, and the fasteners closest to the cantilever span will 

experience greater shear load. One example of this condition sometimes occurs 

with balcony construction in residential buildings; failure to consider the effect 

discussed above has been associated with some notable balcony collapses.

For wood members bolted to concrete, the design lateral values are 

provided in NDS•table 8.2E. The yield equations (or general dowel equations) 

may also be used to conservatively determine the joint capacity. A recent study 

has made recommendations regarding reasonable assumptions that must be made 

in applying the yield equations to bolted wood-to-concrete connections (Stieda, 

1999). Using symbols defined in the NDS, the study recommends an Re value of 5 

and an Rt value of 3. These assumptions are reported as being conservative 

because fastener head effects and joint friction are ignored in the general dowel 

equations.
Residential Structural Design Guide 7-19

7.3.5 Lag Screws

Lag screws (or lag bolts) may be designed to resist shear and withdrawal 

loads in wood-to-wood and metal-to-wood connections in accordance with 

NDS•9. As mentioned, many specialty “screw-type” fasteners can be installed in 

wood. Some tap their own holes and do not require predrilling. Manufacturer data 

should be consulted for connection designs that use proprietary fastening systems.

The withdrawal strength of a lag screw (inserted into the side grain of 

lumber) is determined in accordance with either the empirical design equation 

below or NDS•table 9.2A. It should be noted that the equation below is based on 

single lag screw connection tests and is associated with a reduction factor of 0.2 

applied to average ultimate withdrawal capacity to adjust for load duration and 

safety. Also, the penetration length of the lag screw Lp into the main member does 

not include the tapered portion at the point. NDS•appendix L contains dimensions 

for lag screws.

NDS•9.2.1

 unadjusted withdrawal design value (lb) for a lag screw.

where,

G = specific gravity of the lumber receiving the lag screw tip.

D = the diameter of the lag screw shank (in).

Lp = the depth of penetration (in) of the lag screw into the member receiving the 

tip, less the tapered length of the tip.

The allowable withdrawal design strength of a lag screw is greater when 

the screw is installed in the side rather than the end grain of a member. However, 

unlike the treatment of nails, the withdrawal strength of lag screws installed in the 

end grain may be calculated by using the Ceg adjustment factor with the equation 

above.

The design shear value Z for a lag screw is typically determined by using 

the following tables from NDS•9:

• Table 9.3A. Lag screw, single-shear (two-member) connections with 

the same species of lumber for both members.

• Table 9.3B. Lag screw and metal plate-to-wood connections.

The yield equations in NDS•9.3 may be used for conditions not 

represented in the design value tables for Z. Regardless of the method used to 

determine the Z value for a single lag screw, the value must be adjusted as 

described in section 7.3.2.

Also worth mentioning is that the NDS provides an equation for 

determining the allowable shear design value when a lag screw connection is 

loaded in combined withdrawal and shear (see NDS•9.3.5, equation 9.3-6). The 

p

4

3

2

3

W = 1800(G) D L
Residential Structural Design Guide 7-20

equation does not, however, seem to apply to typical uses of lag screws in 

residential construction.

7.3.6 System Design Considerations

As with any building code or design specification, the NDS provisions 

may not address all conditions encountered in the field. Earlier chapters made 

several recommendations regarding alternative or improved design approaches. 

Similarly, some considerations regarding wood connection design are in order.

First, as a general design consideration, “crowded” connections should be 

avoided. If too many fasteners are used (particularly nails), they may cause 

splitting during installation. When connections become crowded, an alternative 

fastener or connection detail should be considered. Basically, the connection 

detail should be practical and efficient.

Second, although the NDS addresses “system effects” within a particular 

joint (that is, element) that uses multiple bolts or lag screws (that is, the group 

action factor Cg), it does not include provisions regarding the system effects of 

multiple joints in an assembly or system of components. Some consideration of

system effects is given below, therefore, based on several relevant studies related 

to key connections in a home that allow the dwelling to perform effectively as a 

structural unit.

Sheathing Withdrawal Connections

Several studies have focused on roof sheathing attachment and nail 

withdrawal, primarily as a result of Hurricane Andrew (Cunningham, 1993; HUD, 

1999b; McLain, 1997; Mizzell and Schiff, 1994; and Murphy, Pye, and 

Rosowsky, 1995). The studies identify problems related to predicting the pull-off 

capacity of sheathing based on single nail withdrawal values and determining the 

tributary withdrawal load (that is, wind suction pressure) on a particular sheathing 

fastener. One clear finding, however, is that the nails on the interior of the roof 

sheathing panels are the critical fasteners (that is, they initiate panel withdrawal 

failure) because of the generally larger tributary area served by these fasteners. 

The studies also identified benefits to the use of screws and deformed shank nails; 

however, use of a standard geometric tributary area of the sheathing fastener and 

the wind loads, as described in chapter 3, along with the NDS withdrawal values 

(section 7.3.3), will generally result in a reasonable design using nails. The wind 

load duration factor should also be applied to adjust the withdrawal values 

because a commensurate reduction is implicit in the design withdrawal values 

relative to the short-term, tested, ultimate withdrawal capacities.

One study, however, found that the lower bound (that is, fifth percentile) 

sheathing pull-off resistance was considerably higher than that predicted by use of 

single-nail test values (Murphy, Pye, and Rosowsky, 1995). The difference was as 

large as a factor of 1.39 greater than the single nail values. Although this finding 

would suggest a withdrawal system factor of at least 1.3 for sheathing nails, it 

should be subject to additional considerations. For example, sheathing nails are 
Residential Structural Design Guide 7-21

placed by people using tools in somewhat adverse conditions (for example, on a 

roof), not in a laboratory; therefore, this system effect may be best considered as a 

reasonable “construction tolerance” on actual nail spacing variation relative to 

that intended by design. Thus, an 8- to 9-inch nail spacing on roof sheathing nails 

in the panel’s field could be “tolerated” when a 6-inch spacing is “targeted” by 

design.

Roof-to-Wall Connections

Several studies (Conner, Gromala, and Burgess, 1987; Reed, Rosowsky, 

and Schiff, 1996) have investigated the capacity of roof-to-wall (that is, sloped 

rafter to top plate) connections using conventional toenailing and other 

enhancements (strapping, brackets, and gluing, for example). Again, the primary 

concern is related to high wind conditions, such as those experienced during 

hurricane wind events.

First, as a matter of clarification, the toenail reduction factor Ctn does not 

apply to slant nailing, such as in rafter-to-wall and floor-to-wall connections in 

conventional residential construction (Hoyle and Woeste, 1989). Toenailing 

occurs when a nail is driven at an angle in a direction parallel-to-grain at the end 

of a member (for example, a wall stud toenail connection to the top or bottom 

plate that may be used instead of end nailing). Slant nailing occurs when a nail is 

driven at an angle, but in a direction perpendicular-to-grain through the side of the 

member and into the face grain of the other (for example, from a roof rafter or 

floor band joist to a wall top plate). Although a generally reliable connection in 

most homes and similar structures built in the United States, even a well-designed 

slant-nail connection used to attach roofs to walls will become impractical in 

hurricane-prone regions or similar high-wind areas. In these conditions, a metal 

strap, bracket, or clip is required to resist the high uplift loads induced by these 

high-wind events.

Based on the studies of roof-to-wall connections, five key findings are 

summarized, as follows (Conner, Gromala, and Burgess, 1987; Reed, Rosowsky, 

and Schiff, 1996)—

1. In general, studies found that slant nails (not to be confused with toe 

nails), in combination with metal straps or brackets, do not provide 

directly additive uplift resistance.

2. A basic metal twist strap placed on the interior side of the walls (that 

is, the gypsum board side) resulted in top plate tear-out and premature 

failure. A strap placed on the outside of the wall (that is, the structural 

sheathing side), however, was able to develop its full capacity without 

additional enhancement of the conventional stud-to-top plate 

connection.

3. The withdrawal capacity for single joints with slant nails was 

reasonably predicted by NDS with a safety factor of about 2 to 3.5; 

however, with multiple joints tested simultaneously, a system factor on 

withdrawal capacity of greater than 1.3 was found for the slant-nailed 
Residential Structural Design Guide 7-22

rafter-to-wall connection. A similar system effect was not found on 

strap connections, although the strap capacity was substantially higher. 

The ultimate capacity of the simple strap connection (using five 8d 

nails on either side of the strap—five in the spruce rafter and five in 

the Southern Yellow Pine top plate) was found to be about 1,900 

pounds per connection. The capacity of three 8d common slant nails 

used in the same joint configuration was found to be 420 pounds on 

average, and with higher variation. 

4. The strap manufacturer’s published value had an excessive safety 

margin of greater than 5 relative to average ultimate capacity. 

Adjusted to an appropriate safety factor in the range of 2 to 3 (as 

calculated by applying NDS nail shear equations by using a metal side 

plate), the strap (a simple 18g twist strap) would cover a multitude of 

high wind conditions with a simple, economical connection detail.

5. The use of deformed shank (that is, annular ring) nails driven in using 

the slant nailing method was found to increase dramatically the uplift 

capacity of the roof-to-wall connections.

Heel Joint in Rafter-to-Ceiling Joist Connections

The heel joint connection at the intersection of rafters and ceiling joists 

has long been considered one of the weaker connections in conventional wood 

roof framing. In fact, this highly stressed joint is one of the accolades of using a 

wood truss rather than conventional rafter framing (particularly in high-wind or 

snow-load conditions). The designer should understand the performance of 

conventional rafter-to-ceiling joist heel joint connections, however, because they 

are frequently encountered in residential construction.

Conventional rafter and ceiling joist (cross-tie) framing can be analyzed 

by using methods that are applicable to trusses (that is, pinned joint analysis); 

however, the performance of the system should be considered. As mentioned 

earlier for roof trusses (section 5.6.1 in chapter 5), a system factor of 1.1 is 

applicable to tension members and connections; therefore, the calculated shear 

capacity of the nails in the heel joint (and in ceiling joist splices) may be 

multiplied by a system factor of 1.1, which is considered conservative. Second, 

the designer must remember that the nail shear values are based on a deformation 

limit and generally have a conservative safety factor of 3 to 5 relative to the 

ultimate capacity. Finally, the nail values should be adjusted for duration of load 

(that is, a snow load duration factor of 1.15 to 1.25); refer to section 5.2.4 of 

chapter 5. With these considerations and with the use of rafter support braces at or 

near mid-span (as is common), reasonable heel joint designs should be possible 

for most typical design conditions in residential construction.

Wall-to-Floor Connections

Many nails often are used to connect wood sole plates to wood floors, 

particularly along the total length of the sole plate or wall bottom plate. Usually, 
Residential Structural Design Guide 7-23

several bolts along the length of the bottom plate connect it to a concrete slab or 

foundation wall. This points toward the question of possible system effects in 

estimating the shear capacity (and uplift capacity) of these connections for design 

purposes.

In recent shear wall tests, walls connected with pneumatic nails (0.131-

inch diameter by 3 inches long), spaced in pairs at 16 inches on center along the 

bottom plate, were found to resist more than 600 pounds in shear per nail (HUD, 

1999a) (the bottom plate was Spruce-Pine-Fir lumber, and the base beam was 

Southern Yellow Pine). That value is about 4.5 times the adjusted allowable 

design shear capacity predicted using the NDS equations. Similarly, connections 

using 5/8-inch-diameter anchor bolts at 6 feet on center (all other conditions 

equal) were tested in full shear wall assemblies; the ultimate shear capacity per 

bolt was found to be 4,400 pounds. That value is about 3.5 times the adjusted 

allowable design shear capacity per the NDS equations. The designer should 

consider those safety margins when evaluating similar connections from a 

practical “system” standpoint; however, compliance with the applicable standards 

and accepted methodology is still the recommended approach.

7.4 Design of Concrete and Masonry 

Connections

7.4.1 General

In typical residential construction, the interconnection of concrete and 

masonry elements or systems is generally related to the foundation and usually 

handled in accordance with standard or accepted practice. The bolted wood 

member connections to concrete, as described in section 7.3.4, are suitable for 

bolted wood connections to properly grouted masonry (see chapter 4). Moreover, 

numerous specialty fasteners or connectors (including power driven and cast-inplace) can be used to fasten wood materials to masonry or concrete. The designer 

should consult the manufacturer’s literature for available connectors, fasteners, 

and design values.

This section discusses some typical concrete and masonry connection 

designs in accordance with the ACI 318 concrete design specification and ACI

530 masonry design specification (ACI, 2008, 2011).
Residential Structural Design Guide 7-24

7.4.2 Concrete or Masonry Foundation Wall to Footing

Footing connections, if any, are intended to transfer shear loads from the 

wall to the footing below. The shear loads generally are produced by lateral soil 

pressure acting on the foundation (see chapter 3) or loads in shear walls from 

wind or seismic events.

Footing-to-wall connections for residential construction often are 

constructed in one of the following three ways (refer to figure 7.5 for illustrations 

of the connections)—

• No vertical reinforcement or key.

• Key only.

• Dowel only.

Generally, no special connection is needed in non-hurricane-prone or lowto moderate-hazard seismic areas. Friction is sufficient for low, unbalanced 

backfill heights, whereas the basement slab can resist slippage for higher backfill 

heights on basement walls. The basement slab abuts the basement wall near its 

base and thus provides lateral support. If gravel footings are used, the unbalanced 

backfill height must be sufficiently low (that is, less than 3 feet), or means must 

be provided to prevent the foundation wall from slipping sideways from lateral 

soil loads. Again, a basement slab can provide the needed support. Alternatively, 

a footing key or doweled connection can be used. In spite of the many probable 

locations across the country where no physical connection might be required 

between the foundation wall and the footing, the best practice remains to attach 

these two elements. Without the physical connection, this critical part of the load 

path arguably is not continuous. Often, problems arise that could have been 

reduced or eliminated if a physical connection had existed between the foundation 

wall and footing. This is particularly true when any extreme loading occurs, such 

as high winds or even a moderate earthquake.
Residential Structural Design Guide 7-25

FIGURE 7.5 Concrete or Masonry Wall-to-Footing Connections

Friction Used To Provide Shear Transfer

To verify the amount of shear resistance provided by friction alone, the 

designer should assume a coefficient of friction between two concrete surfaces of 

μ = 0.6. Using dead loads only, determine the static friction force, , where 

F is the friction force (pounds), N is the dead load (pounds per square foot), and A

is the bearing surface area (square feet) between the wall and the footing. The 

appropriate load combination must be used in determining the resistance 

permitted for the dead load.

Key Used To Provide Shear Transfer

A concrete key is commonly used to “interlock” foundation walls to 

footings. If foundation walls are constructed of masonry, the first course of 

masonry must be grouted solid when a key is used.

In residential construction, builders often form a key by using a 2x4 wood 

board with chamfered edges and placing it into the surface of the footing 

immediately after the concrete pour. Figure 7.6 illustrates a footing with a key; 

shear resistance developed by the key is computed in accordance with the 

following equation.

ACI-318•22.5

.

F = μ NA

Vu ≤ φVn
Residential Structural Design Guide 7-26

ACI-318•22.5

.

FIGURE 7.6 Key in Concrete Footings

Dowels Used To Provide Adequate Shear Transfer

Shear forces at the base of exterior foundation walls may require a dowel 

to transfer the forces from the wall to the footing. The designer can use the 

equations that follow, described by ACI-318 as the Shear-Friction Method, to 

develop shear resistance with vertical reinforcement (dowels) across the wallfooting interface. 

ACI-318•11.7

Masonry Walls Concrete Walls

φ = 0.85

If dowels are used to transfer shear forces from the base of the wall to the 

footing, the designer can use the equations below to determine the minimum 

f bh

3

4 Vn c = ′

be 12db l ≥ Vu ≤ φVn



















 ′

=

v y

4 m v

v

0.12A f

350 f A

B minimumof 













 ′ = μ ≤

c

c c

n vf y 800A

0.2f A

V A f

φ μ =

y

u vf f

V A
Residential Structural Design Guide 7-27

development length required (refer to figure 7.7 for typical dowel placement). If 

development length exceeds the footing thickness, the dowel must be in the form 

of a hook, which usually is not required in residential construction except in highwind or high-seismic regions or where the foundation wall will resist deep 

backfill or will likely be subject to hydrostatic pressure.

ACI-318•12.2 & 12.5 Concrete Walls

Standard Hooks Deformed Bars 

 where fy = 60,000 psi

ACI-530•1.12.3 & 2.1.8 Masonry Walls

Standard Hooks

FIGURE 7.7 Dowel Placement in Concrete Footings

The minimum embedment length is a limit specified in ACI-318 that is not 

necessarily compatible with residential construction conditions and practice; 

therefore, this guide suggests a minimum embedment length of 6 to 8 inches for 

c

b

hb f

1200d l ′ =





























+

αβγλ

















′ =

b

TR c

y

db

d

40 f c K

3f

l

2.5

d

c K

b

TR ≤

+

60,000

f y ξ = 12

A

A

l l

s,provided

s,required

d db ≥ ′′ 













 =

s,provided

s,required

A

A

ω =

l 0.0015d F 12 in. d = b s ≥

e 25db l = 11.
Residential Structural Design Guide 7-28

footing dowels, when necessary, in residential construction applications. In 

addition, dowels sometimes are used in residential construction to connect other 

concrete elements, such as porch slabs or stairs, to the house foundation to control 

differential movement. Exterior concrete “flat work” adjacent to a home, 

however, should be founded on adequate soil bearing or reasonably compacted 

backfill. Finally, connecting exterior concrete work to the house foundation 

requires caution, particularly in colder climates and soil conditions where frost 

heave may be a concern.

7.4.3 Anchorage and Bearing on Foundation Walls

In addition to the references noted herein, appendix D of ACI 318 is 

devoted to anchoring to concrete and provides substantial information for the 

designer on both shear and tension designs for anchors in concrete.

Anchorage Tension (Uplift) Capacity

The equations that follow determine whether the concrete or masonry 

shear area of each bolt is sufficient to resist pull-out from the wall as a result of 

uplift forces and shear friction in the concrete.

ACI-318•11.3 & ACI-530•2.1.2

Concrete Foundation Wall Masonry Foundation Wall

Bearing Strength

Determining the adequacy of the bearing strength of a foundation wall 

follows ACI-318•10.17 for concrete or ACI-530•2.1.7 for masonry. The bearing 

strength of the foundation wall typically is adequate for the loads encountered in 

residential construction.

ACI-318•10.17 & ACI-530•2.1.7

Concrete Foundation Wall Masonry Foundation Wall

Bc = factored bearing load

φ = 0.7

When the foundation wall’s supporting surface is wider on all sides than 

the loaded area, the designer may determine the design bearing strength on the 

loaded area using the following equations.

Vu ≤ φVc ba ≤ Ba

c v c V = 4A f ′



















 ′

=

b y

p m

a 0.2A f

0.5A f

B minimumof

c cA1 B ≤ φ0.85f ′ a Fa f ≤

1

a A

P f =

Fa 0.25f m ≤ ′
Residential Structural Design Guide 7-29

ACI-318•10.7 & ACI-530•2.1.7

Concrete Foundation Wall

Masonry Foundation Wall

2

A

A where

A

A

B 0.85f A

1

2

1

2 c = φ c

′ 1 ≤

2

A

A where

A A A

P f

1

2

1 2 1

a = ≤
Residential Structural Design Guide 7-30

7.5 Design Examples

EXAMPLE 7.1 Roof Sheathing Connection

Given • Design wind speed is 130 mph gust with an open (coastal) exposure.

• Two-story home with a gable roof

• Roof framing lumber is Southern Yellow Pine (G=0.55)

• Roof framing is spaced at 24 inches on center

• Roof sheathing is 7/16-inch-thick structural wood panel

Find Wind load (suction) on roof sheathing.

Nail type/size and maximum spacing.

Solution

1. Determine the wind load on roof sheathing (chapter 3, section 3.6.2)

Step 1: Basic velocity pressure = 26 psf (table 3.8)

Step 2: Adjust for open exposure = 1.4(26 psf) = 36.4 psf

Step 3: Skip

Step 4: Roof sheathing Gcp = -2.2 (table 3.10)

Step 5: Design load = (-2.2)(36.4 psf) = 80 psf

2. Select a trial nail type and size, determine withdrawal capacity, and calculate 

required spacing.

Use an 8d pneumatic nail (0.113 inch diameter) with a length of 2 3/8 inches. The 

unadjusted design withdrawal capacity is determined using the equation in section 

7.3.3.

W = 1380(G)2.5DLp

G = 0.55

D = 0.113 in

Lp = (2 3/8 in) – (7/16 in) = 1.9 in

W = 1380(0.55)2.5(0.113 in)(1.9 in) = 66.5 lb

Determine the adjusted design withdrawal capacity using the applicable adjustment 

factors discussed in section 7.3.2.

W' = WCD = (66.5 lb)(1.6) = 106 lb

Determine the required nail spacing in the roof sheathing panel interior.

Tributary sheathing area = (roof framing spacing)(nail spacing)

= (2 ft)(s)

Withdrawal load per nail = (wind uplift pressure)(2 ft)(s)

= (80 psf)(2 ft)(s)
Residential Structural Design Guide 7-31

W' ≥ design withdrawal load

106 lb ≥ (80 psf)(2 ft)(s)

s ≤ 0.66 ft

Use a maximum nail spacing of 6 inches in the roof sheathing panel interior.

Notes:

1. If Spruce-Pine-Fir (G=0.42) roof framing lumber is substituted, W' would be 

54 lb, and the required nail spacing would reduce to 4 inches on center in the 

roof sheathing panel interior. Thus, carefully considering and verifying the 

species of framing lumber is extremely important when determining fastening 

requirements for roof sheathing.

2. The preceding analysis is based on a smooth shank nail. A ring shank nail may 

be used to provide greater withdrawal capacity that also is less susceptible to 

lumber moisture conditions at installation and related long-term effects on 

withdrawal capacity.

3. With the smaller tributary area, the roof sheathing edges that are supported on 

framing members may be fastened at the standard 6-inch on center fastener 

spacing. For simplicity, the designer may want to specify a 6-inch on center 

spacing for all roof sheathing fasteners. 

4. As an added measure, given the extreme wind environment, the sheathing nail 

spacing along the gable end truss or framing should be specified at a closer 

spacing, say, 4 inches on center. These fasteners are critical to the performance 

of light-frame gable roofs in extreme wind events; refer to the discussion on 

hurricanes in chapter 1. NDS•12.3.8 provides an equation to determine nail 

lateral strength when subjected to a combined lateral and withdrawal load. 

This equation may be used to verify the 4-inch nail spacing recommendation at 

the gable end.

Conclusion This example problem demonstrates a simple application of the nail withdrawal 

equation in the NDS. The withdrawal forces on connections in residential 

construction usually are of greatest concern in the roof sheathing attachment. In 

hurricane-prone regions, common practice is to use 6-inch nail spacing on the 

interior of roof sheathing panels. 
Residential Structural Design Guide 7-32

EXAMPLE 7.2 Roof-to-Wall Connections

Given • Design wind speed is 120 mph gust with an open coastal exposure.

• One-story home with a hip roof (28 ft clear span trusses with 2 ft overhangs).

• Roof slope is 6:12.

• Trusses are spaced at 24 in on center.

Find 1. Uplift and transverse shear load at the roof-to-wall connection.

2. Connection detail to resist the design loads.

Solution

1. Determine the design loads on the connection (chapter 3).

Dead load (section 3.3)

Roof dead load = 15 psf (table 3.2)

Dead load on wall = (15 psf)[0.5(28 ft) + 2 ft] = 240 plf

Wind load (section 3.6)

Step 1: Basic velocity pressure = 18 psf (table 3.8)

Step 2: Adjust for open exposure = 1.4(18 psf) = 25.2 psf

Step 3: Skip

Step 4: Roof uplift Gcp = -0.8

Overhang Gcp = +0.8

Step 5: Roof uplift pressure = -0.8(25.2 psf) = -20 psf 

Overhang pressure = 0.8 (25.2 psf) = 20 psf 

Determine the wind uplift load on the wall.

Design load on wall = 0.6D + 0.6Wu (table 3.1)

= 0.6 (240 plf) + 0.6{(-20 psf)[0.5(28 ft) + 2 ft] – (20 psf)(2 

ft)}

= - 72 plf (upward)

Design load per wall-to-truss connection = (2 ft)(-72 plf) = -144 lb (upward)

Determine the transverse shear (lateral) load on the roof-to-wall connection. The 

transverse load is associated with the role of the roof diaphragm in supporting and 

transferring lateral loads from direct wind pressure on the walls.

Design lateral load on the wall-to-truss connection

= 1/2 (wall height)(wall pressure)(truss spacing)

Adjusted velocity pressure = 25.2 psf

Wall GCp = -1.2,+1.1*

Wind pressure = 1.1(25.2 psf) = 28 psf

*The 1.1 coefficient is used because the maximum uplift on the roof and roof 

overhang occurs on a windward side of the building (that is, positive wall 

pressure).
Residential Structural Design Guide 7-33

= 1/2 (8 ft)(28 psf)(2 ft) = 224 lb

Thus, roof-to-wall connection combined design loads are as follows.

144 lb (uplift)

224 lb (lateral, perpendicular to wall)*

*The lateral load parallel to a wall is not considered to be significant in this 

example problem, although it may be checked to verify the transfer of lateral wind 

loads on the roof to shear walls; refer to chapter 6.

2. Determine a roof-to-wall connection detail to resist the combined design load.

Generally, manufacturers publish loading data for metal connectors for multiple 

loading directions. The designer should verify that these values are for 

simultaneous, multidirectional loading or make reasonable adjustments as needed. 

In this example problem, the NDS will be used to design a simple roof tie-down 

strap and slant nail connection. A tie-down strap will be used to resist the uplift 

load, and typical slant nailing will be used to resist the lateral load. The slant 

nailing, however, does not contribute appreciably to the uplift capacity when a 

strap or metal connector is used; refer to section 7.3.6.

Uplift load resistance

Assuming an 18g (minimum 0.043 inches thick) metal strap is used, determine the 

number of 6d common nails required to connect the strap to the truss and to the 

wall top plate to resist the design uplift load.

The nail shear capacity is determined as follows.

Z = 60 lb (NDS table 12.3F)

Z' = ZCD (section 7.3.2)

= (60 lb)(1.6)

= 96 lb

The number of nails required in each end of the strap is—

(144 lb)/(96 lb/nail) = 2 nails

The above Z value for metal side-plates implicitly addresses failure modes that 

may be associated with strap/nail head tear-through; however, the width of the 

strap must be calculated. Assuming a minimum 33 ksi steel yield strength and a 

standard 0.6 safety factor, the width of the strap is determined as follows.

0.6(33,000 psi)(0.043 in)(w) = 144 lb

w = 0.17 in

Therefore, use a minimum 1/2-inch-wide strap to allow for the width of nail holes 

and a staggered nail pattern.

Lateral load resistance

Assuming that a 16d pneumatic nail will be used (0.131 in diameter by 3.5 inches 

long), determine the number of slant-driven nails required to transfer the lateral 
Residential Structural Design Guide 7-34

load from the wall to the roof sheathing diaphragm through the roof trusses. 

Assume that the wall framing is Spruce-Pine-Fir (G = 0.42).

Z = 88 lb (NDS table 12.3A)*

*A 1 1/4-inch side member thickness is used to account for the 

slant nail penetration through the truss at an angle.

Z' = ZCD** **The Ctn value of 0.83 is not used because the nail is slant 

driven and is not a toe nail; refer to section 7.3.6.

Z' = (88 lb)(1.6) = 141 lb

Therefore, the number of nails required to transfer the transverse shear load is 

determined as follows.

(224 lb)/(141 lb/nail) = 2 nails

Conclusion The beginning of the uplift load path is on the roof sheathing, which is transferred 

to the roof framing through the sheathing nails; refer to example 7.1. The uplift 

load is then passed through the roof-to-wall connections, as demonstrated in this 

example problem. The designer should note that the load path for wind uplift 

cannot overlook any joint in the framing.

One common error is to attach small roof tie-straps or clips to only the top member 

of the wall top plate. Thus, the uplift load must be transferred between the two 

members of the double top plate, which usually are only face-nailed together for 

assembly, not to transfer large uplift loads. This circumstance would not typically 

be a problem if the wall sheathing were attached to the top member of the double 

top plate, but walls usually are built to an 8 ft-1 in height to allow for assembly of 

interior finishes and to result in a full 8 ft ceiling height after floor and ceiling 

finishes. Because sheathing is nominally 8 ft in length, it cannot span the full wall 

height and may not be attached to the top member of the top plate. Also, the strap 

should be placed on the structural sheathing side of the wall unless framing joints 

within the wall (that is, stud-to-plates) are adequately reinforced.

Longer sheathing can be special ordered and is often used to transfer uplift and 

shear loads across floor levels by lapping the sheathing over the floor framing to 

the wall below. The sheathing may also be laced at the floor band joist to transfer 

uplift load, but the cross-grain tension of the band joist should not exceed a 

suitably low stress value (that is, 1/3Fv); refer to chapter 5, section 5.3.1.
Residential Structural Design Guide 7-35

EXAMPLE 7.3 Wall Sole Plate to Floor Connection

Given • A 2x4 wall bottom (sole) plate of Spruce-Pine-Fir is fastened to a wood floor 

deck.

• Floor framing lumber is Hem-Fir.

• A 3/4-inch-thick wood structural panel subfloor is used.

• The bottom plate is subject to the following design loads because of wind or 

earthquake lateral loads—

o 250 plf shear parallel-to-grain (shear wall slip resistance).

o 120 plf shear perpendicular-to-grain (transverse load on wall).

• The uplift load on the wall, if any, is assumed to be resisted by other 

connections (for example, uplift straps, shear wall hold-downs).

Find A suitable nailing schedule for the wall sole plate connection using 16d pneumatic 

nails (0.131-inch diameter by 3.5 inches long).

Solution

The designer should assume that the nails will penetrate the subflooring and the 

floor framing members. The designer should also conservatively assume that the 

density of the subfloor sheathing and the floor framing is the same as the wall 

bottom plate (lowest density of the connected materials). These assumptions allow 

for the use of NDS table 12.3A. Alternatively, a more accurate nail design lateral 

capacity may be calculated using the yield equations of NDS•12.3.1.

Using NDS table 12.3A, the designer should note that the closest nail diameters in 

the table are 0.135 and 0.128 inches. Interpolating between these values, using a 

side member thickness of 1.5 inches, and assuming Spruce-Pine-Fir for all 

members, the following Z value is obtained—

Z = 79 + [(0.131-0.128)/(0.135-0.128)](88 lb – 79 lb) = 83 lb*

Z' = ZCD = 83 lb (1.6) = 133 lb

*Using the NDS general dowel equations as presented in AF&PA Technical Report 

12 (AF&PA, 1999), the calculated value is identical under the same simplifying 

assumptions. A higher design value of 90 pounds may be calculated, however, by 

using only the subfloor sheathing as a side member, with G = 0.5. The ultimate 

capacity is conservatively predicted as 261 pounds.

Assuming that both of the lateral loads act simultaneously at their full design value 

(a conservative assumption), the resultant design load is determined as follows.

Resultant shear load = sqrt[(250plf)2 + (120 plf)2

] = 277 plf

Using the conservative assumptions above, the number of nails per linear foot of 

wall plate is determined as follows.

(277 lb)/(133 lb/nail) = 2.1 nails per foot

Rounding this number, the design recommendation is 2 nails per foot or 3 nails per 

16 inches of wall plate.
Residential Structural Design Guide 7-36

Conclusion The number of 16d pneumatic nails (0.131 inch diameter) required is 2 nails per 

foot of wall bottom plate for the moderate loading condition evaluated. The number 

of nails may be reduced by using a larger diameter nail or by evaluating the nail 

lateral capacity using the yield equations of NDS•12.3.1.

As in example 7.3, some consideration of extensive experience in conventional 

residential construction should also be considered in view of the conventional 

fastening requirements of table 7.1 for wood sole plate to floor framing connections 

(that is, one 16d nail at 16 inches on center); refer to NDS•7.1.1.4. For the loads 

assumed in this example problem, two nails per 16 inches on center may be 

adequate. Testing has indicated that the ultimate capacity of two 16d pneumatic 

nails (0.131 inch diameter) can exceed 600 lb per nail for conditions similar to 

those assumed in this example problem; refer to section 7.3.6. The general dowel 

equations underpredict the ultimate capacity by approximately a factor of two. 

Using two 16d pneumatic nails at 16 inches on center may be expected to provide a 

safety factor of greater than 3 relative to the design lateral load assumed in this 

problem (that is, [600 lb/nail] x [2nails/1.33 ft]/277 plf = 3.2).

As noted in chapter 6, the ultimate capacity of base connections for shear walls 

should at least exceed the ultimate capacity of the shear wall for seismic design, 

and, for wind design, the connection should provide a safety factor of at least 2 

relative to the wind load. For seismic design, the safety factor for shear walls 

recommended in this guide is 2.5; refer to chapter 6, section 6.5.2.3. Therefore, the 

fastening schedule of two 16d pneumatic nails at 16 inches on center is not quite 

adequate for seismic design loads of the magnitude assumed in this problem (that 

is, the connection does not provide a safety factor of at least 2.5). The reader is 

referred to chapter 3, section 3.8.4 for additional discussion on seismic design 

considerations and the concept of “balanced” design.
Residential Structural Design Guide 7-37

EXAMPLE 7.4 Wood Floor Ledger Connection to a Wood or Concrete Wall

Given • A 3x8 wood ledger board (Douglas-Fir) is used to support a side-bearing floor 

system.

• The ledger is attached to 3x4 wall studs (Douglas-Fir), spaced at 16 inches on 

center in a balloon-framed portion of a home; as a second condition, the ledger 

is attached to a concrete wall.

• The design shear load on the ledger is 300 plf because of floor live and dead 

loads.

Find 1. The spacing of 5/8-inch-diameter lag screws required to fasten the ledger to 

the wood wall framing.

2. The spacing of 5/8-inch-diameter anchor bolts required to fasten the ledger to a 

concrete wall.

Solution

1. Determine connection requirements for use of a 5/8-inch-diameter lag screw.

Z' = ZCDCgC∆Cd (section 7.3.2)

Zs⊥ = 630 lb* (NDS table 9.3A)

CD = 1.0 (normal duration load)

Cg = 0.98 (2 bolts in a row) (NDS table 7.3.6A)

C∆ = 1.0**

Cd = p/(8D) = (3.09 in)/[8(5/8 in)] = 0.62 (NDS•9.3.3)

p = (penetration into main member) – (tapered length of tip of lag screw)***

= 3.5 in – 13/32 in = 3.09 in

*The Zs⊥ value is used for joints when the shear load is perpendicular to the grain 

of the side member (or ledger, in this case).

**A C∆ value of 1.0 is predicated on meeting the minimum edge and end distances 

required for lag screws and bolts; refer to NDS•8.5.3 and NDS•9.4. The required 

edge distance in the side member is 4D from the top of the ledger (loaded edge) 

and 1.5D from the bottom of the ledger (unloaded edge), where D is the diameter 

of the bolt or lag screw. The edge distance of 1.5D is barely met for the nominal 3-

inch-wide (2.5-inch-actual) stud, provided the lag screws are installed along the 

center line of the stud.

***A 6-inch-long lag screw will extend through the side member (2.5 inches thick) 

and penetrate into the main member 3.5 inches. The design penetration into the 

main member must be reduced by the length of the tapered tip on the lag screw (see 

appendix L of NDS for lag screw dimensions).

Z'= (630 lb)(1.0)(0.98)(1.0)(0.62) = 383 lb

The lag bolt spacing is determined as follows.

Spacing = (383 lb/lag screw)/(300 plf) = 1.3 ft

Therefore, one lag screw per stud-ledger intersection may be used (that is, 1.33 ft 

spacing). The lag screws should be staggered about 2 inches from the top and 

bottom of the 3x8 ledger board. Because the bolts are staggered (that is, no two 
Residential Structural Design Guide 7-38

bolts in a row), the value of Cg may be revised to 1.0 in the preceding calculations.

2. Determine connection requirements for use of a 5/8-inch-diameter anchor bolt in a 

concrete wall.

Z' = ZCDCgC∆ (section 7.3.2)

Z⊥ = 650 lb* (NDS table 8.2E)

CD = 1.0 (normal duration load)

Cg = 1.0** 

C∆ = 1.0***

* The Z⊥ value is used because the ledger is loaded perpendicular to grain.

**The bolts will be spaced and staggered, not placed in a row.

***Edge and end distance requirements of NDS•8.5.3 and NDS•8.5.4 will be met 

for full design value.

Z' = (650 lb)(1.0)(1.0)(1.0) = 650 lb

The required anchor bolt spacing is determined as follows.

Spacing = (650 lb)/(300 plf) = 2.2 ft

Therefore, the anchor bolts should be spaced at approximately 2 ft on center and 

staggered from the top and bottom edge of the ledger by a distance of 

approximately 2 inches.

Note: In conditions in which this connection is also required to support the wall 

laterally (that is, an outward tension load because of seismic loading on a heavy 

concrete wall), the tension forces may dictate additional connectors to transfer the 

load into the floor diaphragm. In lower wind or seismic load conditions, the ledger 

connection to the wall and the floor sheathing connection to the ledger are usually 

sufficient to transfer the design tension loading, even though that situation may 

induce some cross-grain tension forces in the ledger. The cross-grain tension stress 

may be minimized by locating every other bolt as close to the top of the ledger as is 

practical or by using a larger plate washer on the bolts.

Conclusion The design of bolted side-bearing connections was presented in this design 

example for two wall construction conditions. Although not a common connection 

detail in residential framing, it is one that requires careful design consideration and 

installation because it must transfer the floor loads (that is, from people) through a 

shear connection rather than through a simple bearing. The example also addresses 

the issue of appropriate bolt location with respect to edge and end distances. 

Finally, the designer was alerted to special connection detailing considerations in 

high wind and seismic conditions.
Residential Structural Design Guide 7-39

EXAMPLE 7.5 Wood Sill Connection to Foundation Wall

Given • The foundation wall is connected to a wood sill plate and laterally supported, 

as shown in the following figure.

• Assume that the soil has a 30 pcf equivalent fluid density, and the unbalanced 

backfill height is 7.5 ft.

• The foundation wall unsupported height (from basement slab to top of wall) is 

8 ft.

• The wood sill is preservative-treated Southern Yellow Pine.

Find 1. The lateral load on the foundation wall to sill plate connection resulting from 

the backfill lateral pressure.

2. The required spacing of 1/2-inch-diameter anchor bolts in the sill plate.

Solution

1. Determine the lateral load on the sill plate connection.

Using the procedure in section 3.5 of chapter 3 and the associated beam equations 

in appendix A, the reaction at the top of the foundation wall is determined as 

follows.

Rtop = ql3/(6L) = (30 pcf)(7.5 ft)3/[6(8 ft)] = 264 plf

2. Determine the design lateral capacity of the anchor bolt and the required spacing.

Z' = ZCDCMCtCgC∆ (section 7.3.2)

Z⊥ = 400 lbs* (NDS table 8.2E)

CD = 0.9 (lifetime load duration, table 5.3)

CM = 1.0 (MC < 19%)

Ct = 1.0 (temperature < 100o

F)

Cg = 1.0 (bolts not configured in rows)
Residential Structural Design Guide 7-40

*The value is based on a recommended 6-inch standard embedment of the anchor 

bolt into the concrete wall. Based on conventional construction experience, this 

value may also be applied to masonry foundation wall construction when bolts are 

properly grouted into the masonry wall (that is, by use of a bond beam).

Z' = (400 lb)(0.9) = 360 lb

Anchor bolt spacing = (360 lb)/(264 plf) = 1.4 ft

The check for conformance to ACI 318, appendix D, requires a check of shear on 

the bolt and concrete breakout strength caused by shear. 

Bolt shear for hooked bolt anchors is determined by ACI 318 eq. D-20—

[below.]

V = n0.6Af, where:

n = number of bolts (in this example, 1).

A = effective cross-sectional area of bolt, including consideration for 

threads (in this example, assume A = cross-sectional area of 1⁄2" bolt = πd2/4 = 0.2 

in2

).

f = yield stress of anchor to be taken as the lesser of 1.9fy or 125,000 psi. 

Because the bolt yield strength was not provided, assume 125,000 psi is the lesser.

So V = 1*0.6*.2 in2

*125,000 psi = 15,000 lbs. > 264 plf*4 ft (see note) OK

Concrete breakout strength is determined by ACI 318, eq. D-21 for shear 

perpendicular to the edge of concrete. Assume the bolt is in the middle of the 8 in. 

wall.

Vc = (Avc/Avco)*ψed*ψc*ψh*Vbolt, where:

Avc/Avco = ratio of plan area of concrete breakout/volume of concrete breakout.

Ψed = modification factor for edge effect, in this case = 1.0.

Ψc = modification factor for service loads and cracking, in this case = 1.4.

Ψh = modification factor for thin concrete members, in this case = 1.0.

Vbolt = 264 plf*4 ft = 1,056 lbs.

Vc = (24/72)*1*1.4*1*1056 = 493 lbs.

The allowable shear in the concrete wall is given by ACI 318.11.11 and is as 

follows.

V = 4√f'cd, where:

f'c = design compressive strength of concrete, assume for this example

= 3,000 psi.

d = depth of anchor bolt in concrete, in this example = 4 in.

V = 4*54.8*4 = 877 lb > 493 lb OK

Note: According to the above calculations, an anchor bolt spacing of about 16 

inches on center is required in the sill plate. However, in conventional residential 

construction, extensive experience has shown that a typical anchor bolt spacing of 

6 ft on center is adequate for typical conditions, as represented in this design 

example. This conflict between analysis and experience creates a dilemma for the 

designer that may be reconciled only by making judgmental use of the “extensive 
Residential Structural Design Guide 7-41

experience” clause in NDS•7.1.1.4. Perhaps a reasonable compromise would be to 

require the use of a 5/8-inch-diameter anchor bolt at a 4 ft on center spacing. This 

design may be further justified by consideration of friction in the connection (that 

is, a 0.3 friction coefficient with a typical force resulting from dead load of the 

building). The large safety factor in wood connections may also be attributed to 

some of the discrepancy between practice or experience and analysis in accordance 

with the NDS. Finally, the load must be transferred into the floor framing through 

connection of the floor to the sill. In applications in which the loads are anticipated 

to be much greater (that is, a taller foundation wall with heavier soil loads), the 

joint may be reinforced with a metal bracket, as shown.

Conclusion This example demonstrates an analytic method of determining foundation lateral 

loads and the required connections to support the top of the foundation wall 

through a wood sill plate and floor construction. It also demonstrates the 

discrepancy between calculated connection requirements and conventional 

construction experience that may be negotiated by permissible designer judgment 

and use of conventional residential construction requirements.
Residential Structural Design Guide 7-42

EXAMPLE 7.6 Concrete Wall to Footing (Shear) Connection

Given Maximum transverse shear load on bottom of wall is 1,050 plf (from soil).

Dead load on wall is 1,704 plf.

Yield strength of reinforcement is 60,000 psi.

Wall thickness is 8 inches.

Assume μ = 0.6 for concrete placed against hardened concrete not intentionally 

roughened.

f'c = 3,000 psi

Find Whether a dowel or key is required to provide increased shear transfer capacity. (If 

a dowel or key is required, size accordingly.)

Solution

1. Determine factored shear load on wall due to soil load (that is, 1.6H per chapter 3, 

table 3.1).

V = 1,050 plf

Vu = 1.6 (1,050 plf) = 1,680 plf

2. Check friction resistance between the concrete footing and wall.

Vfriction = μN = μ(dead load per foot of wall)

= (0.6)(1,704 plf) = 1,022 plf < Vu = 1,680 plf

Therefore, a dowel or key is needed to secure the foundation wall to the footing.

3. Determine a required dowel size and spacing (section 7.2 and ACI-318•5.14).

Avf = Vu / (φfyμ)

= (1,680 plf)/[(0.85)(60,000)(0.6)] = 0.05 in2 per foot of wall

Try a No. 4 bar (Av = 0.20 in2

) and determine the required dowel spacing as 

follows:

Avf = Av/S

0.05 in2

/lf = (0.2 in2)/S

S = 48 inches

Conclusion This example problem demonstrates that, for the given conditions, a minimum of 

one No. 4 rebar at 48 inches on center is required to adequately restrict the wall 

from slipping. Alternatively, a key may be used or the basement slab may laterally 

support the base of the foundation wall.

It should be noted that the factored shear load due to the soil lateral pressure is 

compared to the estimated friction resistance in step 1 without factoring the friction 

resistance; no clear guideline exists in this matter of designer judgment.
Residential Structural Design Guide 7-43

EXAMPLE 7.7 Concrete Anchor

Given • A 1/2-inch-diameter anchor bolt is 4 feet on center with a 6-inch embedment 

depth in an 8-inch thick concrete wall.

• The bolt is an ASTM A36 bolt with fy = 36 ksi, and the following design 

properties for ASD; refer to AISC Manual of Steel Construction (AISC, 

2011)—

Ft = 19,100 psi (allowable tensile stress)

Fu = 58,000 psi (ultimate tensile stress)

Fv = 10,000 psi (allowable shear stress)

• The specified concrete has f'c = 3,000 psi.

• The nominal design (unfactored) loading conditions are as follows.

Shear load = 116 plf

Uplift load = 285 plf

Dead load = 180 plf

Find Determine whether the bolt and concrete are adequate for the given conditions.

Solution

1. Check shear in bolt using appropriate ASD steel design specifications (AISC, 

2011) and the ASD load combinations in chapter 3.

fv = = 2,367 psi

Fv = 10,000 psi

fv ≤ Fv OK

2. Check tension in bolt resulting from uplift using appropriate ASD steel design 

specifications (AISC, 2011) and the appropriate ASD load combination in chapter 

3. 

T = [ (285 plf) - 0.6 (180 plf)] (4 ft) = 708 lb

ft = = = 3,612 psi

ft ≤ Ft

3,612 psi < 19,100 psf OK

3. Check tension in concrete (anchorage capacity of concrete) using ACI-318•11.3 

(and ACI 318, appendix D) and the appropriate LRFD load combination in chapter 

3. Note that the assumed cone shear failure surface area, Av, is approximated as the 

minimum of π (bolt embedment length)2 or π (wall thickness)2

.

Vu = T = [1.5 (285 plf) - 0.9 (180 plf)] (4 ft) = 1,062 lb

Av = minimum of 

φVc = φ4Av = (0.85)(4)(113 in2

)= 21,044 lb

(0.196 in )

116 plf (4 ft)

bolt area

shearload

2 =

Abolt

T

2 0.196 in

708 lb









π =

π =

2 2

2 2

(8 in) 201 in

(6 in) 113 in

c f'
Residential Structural Design Guide 7-44

Vu ≤φVc

1,062 lb ≤ 21,044 lb OK

Conclusion A 1/2-inch-diameter anchor bolt with a 6-inch concrete embedment and spaced 4 

feet on center is adequate for the given loading conditions. In lieu of using an 

anchor bolt, the designer may choose to use one of the many strap anchors that are 

also available. The strap anchor manufacturer typically lists the embedment length 

and concrete compressive strength required corresponding to strap gauge and shear 

and tension ratings. In this instance, a design is not typically required; the designer 

simply ensures that the design loads do not exceed the strap anchor’s rated 

capacity.
Residential Structural Design Guide 7-45

7.6 References

American Concrete Institute (ACI). 2008. Building Code Requirements for 

Masonry Structures. ACI Standard 530. Farmington Hills, MI: American 

Concrete Institute. 

———. 2011. Building Code Requirements for Structural Concrete and 

Commentary. ACI Standard 318-11. Farmington Hills, MI: American 

Concrete Institute.

American Forest & Paper Association (AF&PA). 1999. General Dowel Equations 

for Calculating Lateral Connection Values. Technical Report 12. 

Washington, DC American Forest & Paper Association.

American Institute of Steel Construction (AISC). 2011. Manual of Steel 

Construction, 14th Ed. Chicago, IL, American Institute of Steel 

Construction .

American Society of Testing and Materials (ASTM). 2011. Standard 

Specification for Driven Fasteners: Nails, Spikes, and Staples, ASTM 

F1667. West Conshohocken, PA: American Society of Testing and 

Materials.

———. 2012. Standard Terminology of Nails for Use With Wood and Wood-Base 

Materials. ASTM F547. West Conshohocken, PA: American Society of 

Testing and Materials.

American Wood Council (AWC). 2012. National Design Specification for Wood 

Construction (NDS). Washington, DC: American Wood Council.

———. 2014. General Dowel Equations for Calculating Lateral Connection 

Values. Technical Report 12. Washington, DC: American Wood Council.

Conner, H., D. Gromala, and D. Burgess. 1987. “Roof Connections in Houses: 

Key to Wind Resistance,” Journal of Structural Engineering 113 (12): 

2459-2474.

Cunningham, T.P. 1993. Roof Sheathing Fastening Schedules for Wind Uplift. 

APA Report T92-28. Tacoma, WA: American Plywood Association.

Hoyle, R.J., and F.R. Woeste. 1989. Wood Technology in the Design of 

Structures, 5th ed. Ames, IA: Iowa State University Press.

U.S. Department of Housing and Urban Development (HUD). 1999a. Perforated 

Shear Walls With Conventional and Innovative Base Restraint 

Connections. Prepared by the National Association of Home Builders 
Residential Structural Design Guide 7-46

(NAHB) Research Center, Inc., for the U.S. Department of Housing and 

Urban Development. Washington, DC: Government Printing Office.

———. 1999b. Reliability of Conventional Residential Construction: An 

Assessment of Roof Component Performance in Hurricane Andrew and 

Typical Wind Regions of the United States. Prepared by the National 

Association of Home Builders (NAHB) Research Center, Inc., for the U.S. 

Department of Housing and Urban Development. Washington, DC: 

Government Printing Office.

International Code Council, Inc. (ICC). 2012. International Residential Code. 

Country Club Hills, IL: International Code Council, Inc.

International Code Council Evaluation Services (ICC-ES). 2013. Power-Driven 

Staples and Nails. Report No. ESR-1539. Whittier, CA: International 

Code Council Evaluation Services.

McLain, T.E. 1997. “Design Axial Withdrawal Strength From Wood: II. PlainShank Common Wire Nails,” Forest Products Journal 47 (6): 103.

Mizzell, D.P., and S.D. Schiff. 1994. Wind Resistance of Sheathing for 

Residential Roofs. Clemson, SC: Clemson University.

Murphy, S., S. Pye, and D. Rosowsky. 1995. System Effects and Uplift Capacity 

of Roof Sheathing Fasteners. New York: Structures Congress, American 

Society of Civil Engineers. 

Reed, T.D., D.V. Rosowsky, and S.D. Schiff. 1996. Uplift Capacity of Rafter-toWall Connections in Light-Frame Construction. Clemson, SC: Clemson 

University.

Showalter, J., P. Line, and P. Douglas. 1999. Calculation of Lateral Connection 

Values Using General Dowel Equations. American Society of Agricultural 

Engineers, Paper No. 994176. St. Joseph, MI: ASAE.

Stieda, C.K.A. 1999. Lateral Strength of Bolted Wood-to-Concrete Connections, 

Vancouver, British Columbia, Canada: Forintek Canada Corp.
Residential Structural Design Guide A-1

Appendix A

Shear and Moment 

Diagrams and

Beam Equations

q = equivalent fluid density of soil (pcf)

qh = soil pressure (psf) at x = 0

V2 = 

V1 = R1 = 

Vx = V1 (where x < h)

Vx = V2 (where x ≥ h)

Mx = V1x (where x < h)

Mx = −V2 (L - x) (where x ≥ h)

Figure A.1. Simple Beam (Foundation Wall)—Partial Triangular Load

6L

qh R

3

2

− − =

) 3L

h qh (l

2

1 2 −

xq (2h x) 2

1 − −

2 3 qx

6

1 qhx

2

1 − +

q

2V x h h 2 1

@ Mmax = − −

















∆ ≅ ≅ − − +

144L

h

48

h

960

L

128

hL

EI

qL ) 2 (at x L

3 2 2 3

max
Residential Structural Design Guide A-2

Vmax = R2 = Mmax

L

M1 = P1e1

M2 = P2e2

Mmax = |M2| − |M1|

where |M2| > |M1|

Mmax = |M1| − |M2|

where |M1| > |M2|

Figure A.2. Simple Beam (Wall or Column)—Eccentric Point Load

Figure A.3. Simple Beam—Uniformly Distributed Load

( )

( ) 3 2 3

x

4

max

x

2

max

x

max

L 2Lx x 24EI

wx

)

2

L (at x 384EI

5wL

L x 2

wx M

)

2

L (at x 8

wL M

x 2

L V w

2

wL R V

∆ = − +

∆ = =

= −

= =





 



 = −

V = =

V

V
Residential Structural Design Guide A-3

Figure A.4. Simple Beam—Load Increasing Uniformly to One End

R1 = V1 (max when a<b) = 

R2 = V2 (max when a>b) = 

Mmax (at point of load) = 

Mx (when x<a) =

∆max [at x = when a<b] = 

∆a (at point of load) = 

∆x (when x<a) = (L2 - b2 - x2

)

Figure A.5. Simple Beam—Concentrated Load at Any Point

L

Pb

L

Pa

L

Pab

L

Pbx

3

a (a + 2b)

27EIL

Fab (a + 2b) 3a (a + 2b)

3EIL

Pa b

2 2

6EIL

Pbx
Residential Structural Design Guide A-4

R1 = V1 = 

R2 = V2 = 

Vx [when a<x < (L - b)] = R1 – P1

M1 (max when R1<P1) = R1a

M2 (max when R2<P2) = R2b

Mx (when x<a) = R1x

Mx [when a<x <(L - b)] = R1x - P1(x - a)

Figure A.6. Simple Beam—Two Unequal Concentrated Loads Unsymmetrically Placed

R = Vmax = wL

Vx = wx

Mmax (at fixed end) = 

Mx = 

∆max (at free end) = 

∆x = (x4 – 4L3

x + 3L4)

Figure A.7. Cantilever Beam—Uniformly Distributed Load

L

P (L a) P b 1 − + 2

L

P a P (L b) 1 + 2 −

2

wL2

2

wx

2

8EI

wL4

24EI

w
Residential Structural Design Guide A-5

R = V = P

Mmax (at fixed end) = Pb

Mx (when x>a) = P(x-a)

∆max (at free end) = (3L - b)

∆a (at point of load) = 

∆x (when x<a) = (3L - 3x - b)

∆x (when x>a) = (3b - L + x)

Figure A.8. Cantilever Beam—Concentrated Load at Any Point

R1 = V1 = 

R2 = V2 = Vmax = 

Vx = R1 - wx

Mmax = 

M1 (at x = ) = 

Mx = R1x -

∆max (at x = = 0.42L) = 

∆x = (L3 – 3Lx2 + 2x3

)

Figure A.9. Beam Fixed at One End, Supported at Other—Uniformly Distributed Load

6EI

Pb2

3EI

Pb3

6EI

Pb2

6EI

P(L x)

2 −

8

3wL

8

5wL

8

wL2

2 wL

128

9 L

8

3 =

2

wx

2

(1 33) 16

L

+

185EI

wL4

48EI

wx
Residential Structural Design Guide A-6

R1 = V1 = (a + 2L)

R2 = V2 = (3L2 - a2

)

M1 (at point of load) = R1a

M2 (at fixed end) = (a + L)

Mx (when x<a) = R1x

Mx (when x>a) = R1x - P(x-a)

∆max (when a<0.4L at x = ) = 

∆max (when a>0.4L at x = ) = 

∆a (at point of load) = (3L + a)

∆x (when x<a) = (3aL2 – 2Lx2 - ax2

)

∆x (when x>a) = (L-x)2

(3L2x - a2x - 2a2L)

Figure A.10. Beam Fixed at One End, Supported at Other—Concentrated Load at Any Point

R = V = 

Vx = 

Mmax (at ends) = 

M1 (at center) = 

Mx = (6Lx - L2 - 6x2

)

∆max (at center) = 

∆x = (L - x)2

Figure A.11. Beam Fixed at Both Ends—Uniformly Distributed Load

3

2

2L

Pb

3 2L

Pa

2 2L

Pab

2 2L

Pab

2 2

2 2

3L a

L a L −

+

3EI

Pa

2 2 2

2 2 3

(3L a )

(L a )

−

−

2L a

a L

+ 2L a

a

6EI

Pab2

+

3

2 3

12EIL

Pa b

3

2

12EIL

Pa x

3 12EIL

Pa

2

wL





 



 − x

2

L

w

12

wL2

24

wL2

12

w

384EI

wL4

24EI

wx

2
Residential Structural Design Guide A-7

R1 = V1 (max. when a<b) = (3a + b)

R2 = V2 (max. when a>b) = (a + 3b)

M1 (max. when a<b) = 

M2 (max. when a>b) = 

Ma (at point of load) = 

Mx (when x<a) = 

∆max (when a>b at x = ) = 

∆a (at point of load) = 

∆x (when x<a) = (3aL - 3ax - bx)

Figure A.12. Beam Fixed at Both Ends—Concentrated Load at Any Point

R1 = V1 = (L2 - a2

)

R2 = V2 + V3 = (L + a)2

V2 = wa

V3 = (L2 + a2

)

Vx (between supports) = R1 - wx

Vx1 (for overhang) = w(a - x1)

M1 (at x = ) = 

M2 (at R2) = 

Mx (between supports) = (L2 - a2 - xL)

Mx1 (for overhang) = (a - x1)

2

supports) = (L4 – 2L2

x2 + Lx3 ∆x (between -

2a2

L2 + 2a2x2)

∆x1 (for overhang) = (4a2

L - L3 + 6a2x1 - 4ax1

2 + x13

)

Figure A.13. Beam Overhanging One Support—Uniformly Distributed Load

3

2

L

Pb

3

2

L

Pa

2

2

L

Pab

2

2

L

Pa b

3

2 2

L

2Pa b

2

2

1 L

Pab R x −

3a b

2aL

+ 2

3 2

3EI(3a b)

2Pa b

+

3

3 3

3EIL

Pa b

3

2 2

6EIL

Pb x

2L

w

2L

w

2L

w















 − 2

2

L

a 1

2

L 2 2

2 (L a) (L a)

8L

w

+ −

2

wa 2

2L

wx

2

w

wx

24EIL

24EI

wx1
Residential Structural Design Guide A-8

R1 = V1 = 

R2 = V1 + V2 = (L + a)

V2 = P

Mmax (at R2) = Pa

Mx (between supports) = 

Mx1 (for overhang) = P(a - x1)

∆max (between supports at x = ) = 

∆max (for overhang at x1 = a) = (L + a)

∆x (between supports) = (L2 - x3

)

∆x (for overhang) = (2aL + 3ax1 - x1

2

)

Figure A.14. Beam Overhanging One Support—Concentrated Load at End of Overhang

R1 = V1 = R3 = V3 = 

R2 = 

V2 = Vm = 

Mmax = 

M1 [at x = ] = 

Mx [at x < L] = 

∆max [at x ≅ 0.46L] = 

Figure A.15 Continuous Beam—Two Equal Spans and Uniformly Distributed Load

L

Pa

L

P

L

Pax

3

L

9 3EI

PaL2

3EI

Pa 2

6EIL

Pax

6EI

Px1

8

3wL

8

10wL

8

5wL

8

wL2

−

8

3L

128

9wL2

2

wx

8

3wLx 2

−

2

wx

8

3wLx 2

−

185EI

wL4
Residential Structural Design Guide A-9

R1 = V1 = 

R2 = V2 + V3 = 

R3 = V3 = 

V2 = 

Mmax [at x = ] = 

M1 [at R2] = 

Mx [at x < L] = (7L - 8x)

∆max [at x ≅ 0.47L] = 

Figure A.16. Continuous Beam—Two Equal Spans With Uniform Load on One Span

R1 = V1 = + 

R2 = wL1 + wL2 - R1 - R3

R3 = V4 = + 

V2 = wL1 - R1

V3 = wL2 - R3

M1 [at x < L1 , max. at x = ] = R1x = 

M2 = 

M3 [at x1 < L2 , max. at x1 = ] = 

R3x1 -

Figure A.17. Continuous Beam—Two Unequal Spans and Uniformly Distributed Load

wL

16

7

wL

8

5

wL

16

1 −

wL

16

9

L

16

7 2 wL

512

49

2 wL

16

1 −

16

wx

16

wx

109EI

wL4

1

1

L

M

2

wL1

1

1

L

M

2

wL2

w

R1

w

R1

2

wx

2

8 (L L )

wL wL

1 2

3

1

3

2

+

+ − 8 (L L )wL wL

1 2

3

1

3

2

+

+ −

w

R3

w

R3

2

wx

2

1
Residential Structural Design Guide B-1

Appendix B

Unit Conversions

The following list provides the conversion relationship between U.S. customary units and the International System 

of Units (SI). A complete guide to SI and its use can be found in ASTM E380, Standard for Metric Practice.

To Convert From to Multiply by

Length

inch (in) centimeter (cm) 2.54

inch (in) meter (m) 0.0254

foot (ft) meter (m) 0.3048

yard (yd) meter (m) 0.9144

mile (mi) kilometer (km) 1.61

Area

square foot (sq ft) square meter (sq m) 0.0929030

square inch (sq in) square centimeter (sq cm) 6.452

square inch (sq in) square meter (sq m) 0.00064516

square yard (sq yd) square meter (sq m) 0.839127

square mile (sq mi) square kilometer (sq km) 2.6

Volume

cubic inch (cu in) cubic centimeter (cu cm) 16.38706

cubic inch (cu in) cubic meter (cu m) 0.00001639

cubic foot (cu ft) cubic meter (cu m) 0.0283168

cubic yard (cu yd) cubic meter (cu m) 0.7645549

gallon (gal) U.S. liquid liter (l) 3.78541

gallon (gal) U.S. liquid cubic meter (cu m) 0.00378541

Force

kip (1,000 lb) kilogram (kg) 453.6

kip (1,000 lb) Newton (N) 4,448.22

pound (lb) kilogram (kg) 0.453592

pound (lb) Newton (N) 4.44822

Stress or pressure

kip/square inch (ksi) megapascal (Mpa) 6.89476

kip/square inch (ksi) kilogram/square centimeter (kg/sq cm) 70.31

pound/square inch (psi) kilogram/square centimeter (kg/sq cm) 0.07031

pound/square inch (psi) pascal (Pa)b 6,894.757

pound/square inch (psi) megapascal (Mpa) 0.00689476

pound/square foot (psf) kilogram/square meter (kg/sq m) 4.8824

pound/square foot (psf) pascal (Pa) 47.88
Residential Structural Design Guide B-2

Mass (weight)

pound (lb) avoirdupois kilogram (kg) 0.453592

ton (2,000 lb) kilogram (kg) 907.185

Mass (weight) per length

kip/linear foot (klf) kilogram per meter (kg/m) 1,488

pound/linear foot (plf) kilogram per meter (kg/m) 1.488

Moment

foot-pound (ft-lb) Newton-meter (N-m) 1.356

Mass per volume (density)

pound per cubic foot (pcf) kilogram per cubic meter (kg/cu m) 16.0185

pound per cubic yard (lb/cu yd) kilogram per cubic meter (kg/cu m) 0.5933

Velocity

mile per hour (mph) kilometer per hour (km/hr) 1.60934

mile per hour (mph) kilometer per second (km/sec) 0.44704

Temperature

degree Fahrenheit (°F) degree Celsius (°C) tC = (tF-32)/1.8

a

A pascal equals 1,000 Newton per square meter.

The following prefixes and symbols are commonly used to form names and symbols of the decimal multiples and 

submultiples of the SI units.

Multiplication Factor Prefix Symbol

1,000,000,000 = 109 giga G

1,000,000 = 106 mega M

1,000 = 103 kilo k

0.01 = 10-2 centi c

0.001 = 10-3 milli m

0.000001 = 10-6 micro μ

0.000000001 = 10-9 nano n
U.S. Department of Housing and Urban Development

Office of Policy Development and Research

Washington, DC 20410-6000

October 2017