Structural Engineering and Mechanics, Vol. 41, No. 1 (2012) 1-24

1

Shake table tests on a non-seismically detailed
RC frame structure
Akanshu Sharma*, G.R. Reddy and K.K. Vaze
Reactor Safety Division, Bhabha Atomic research Centre, Mumbai 400085, India(Received August 4, 2010, Revised July 8, 2011, Accepted October 25, 2011)
Abstract. A reinforced concrete (RC) framed structure detailed according to non-seismic detailingprovisions as per Indian Standard was tested on shake table under dynamic loads. The structure had 3
main storeys and an additional storey to simulate the footing to plinth level. In plan the structure wassymmetric with 2 bays in each direction. In order to optimize the information obtained from the tests,
tests were planned in three different stages. In the first stage, tests were done with masonry infill panels
in one direction to obtain information on the stiffness increase due to addition of infill panels. In second
stage, the infills were removed and tests were conducted on the structure without and with tuned liquid
dampers (TLD) on the roof of the structure to investigate the effect of TLD on seismic response of thestructure. In the third stage, tests were conducted on bare frame structure under biaxial time histories with
gradually increasing peak ground acceleration (PGA) till failure. The simulated earthquakes represented
low, moderate and severe seismic ground motions. The effects of masonry infill panels on dynamic
characteristics of the structure, effectiveness of TLD in reducing the seismic response of structure and thefailure patterns of non-seismically detailed structures, are clearly brought out. Details of design andsimilitude are also discussed.
Keywords: shake table test; structural engineering; RC structure; dynamic loads; masonry infill panels;
tuned liquid dampers; seismic response; nonlinear behaviour; similitude requirements

1. Introduction
Seismic response of reinforced concrete (RC) framed structures has been one of the major topics
of interest for past few years among the structural engineering researchers around the world. The
main reason behind this interest is the complex behaviour of the RC structures and theirvulnerability to earthquakes. Past earthquakes have exposed the vulnerability of RC structures underseismic loads. Two most important and most frequently found deficiencies in old structures are noshear reinforcement in the joint and poor end anchorages leading to joint shear and bond failures
respectively. Often such failures lead to partial or complete collapse of structures. Fig. 1 shows nearcomplete collapse of structures due to joint failures during 2004 Sumatra Earthquake (Saatcioglu etal. 2004).
Over the years, researchers have investigated the behavior of RC structures under earthquakes andupdated the design guidelines so that the structural performance during earthquakes can be improved.
*Corresponding author, Scientific Officer, E-mail: akanshusharma@yahoo.co.in, akanshu@barc.gov.in

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Akanshu Sharma, G.R. Reddy and K.K. Vaze

Fig. 1 Joint failures causing partial to total collapse of structures

However, there are numerous buildings around the world built without considering such provisions.
In order to make them seismically acceptable, good models are required to evaluate the existingstate of the structure against seismic forces and also to develop various retrofit strategies. Realistic
shake table tests are the best possible way to generate information on real-life behavior of structuresthat can be used to validate the existing or proposed models as well as to evaluate the efficacy of a
retrofit solution. In this work, such an attempt to conduct shake table tests on a reinforced concrete
framed structure that was designed following the non-seismic provisions (IS 456:2000) and generate
useful information is made. The tests were conducted in three different stages as explained below.
One of the most important parameter that needs to be estimated correctly for better prediction offorces that will be attracted by the structure is its fundamental frequency. It is well known that thenatural frequency of RC structures is largely dependent on whether it has masonry infill panels or
not. Infill panels are known to contribute significantly towards the stiffness of the structure. In orderto obtain the effect of masonry infill panels on the stiffness of the structure, sine sweep tests wereconducted on the structure with and without infill panels. However, it must be noted that this was
not the main objective of the tests and was included as the first stage of the program to get
additional information.
Often, it is found that old structures need retrofitting to qualify for the current seismic requirements.
There are numerous such retrofitting techniques now available and many more are continuously
being developed. However, most of the conventional retrofit methods are quite invasive and theyrequire lot of alterations to the original structure. One retrofit solution which is extremely low
invasive and virtually does not disturb any of the inhabitants is the Tuned Liquid Dampers (TLD).
Preliminary tests on a small steel structure (Sharma et al. 2008) provided promising results thatencouraged the authors to try the system on this structure under realistic ground motions. Although,TLD were initially used to control the wind response of the structures (Soong 1988, Modi et al.
1990, Fujino et al. 1992), more recently, the use of TLD in controlling the seismic response ofstructures is studied (Reed et al. 1998, Banerji et al. 2000, Banerji 2004, Li et al. 2004, Jin et al.
2007, Bairrao 2008). It has been found that if properly proportioned, TLD can be quite effective incontrolling the seismic response of structures, even to broadband excitations (Banerji 2004). In thesecond stage of this experimental program, tests were carried on the 3D RC framed structure
without and with TLD and the seismic response were studied and compared.

Shake table tests on a non-seismically detailed RC frame structure

3

Inelastic behaviour of structures plays a vital role in the current seismic design and re-qualification
practice for RC structures. In this practice, the estimated forces on the structure obtained by thelinear analysis are reduced to account for nonlinearity in the structure, or sometimes some kind of
nonlinear analysis such as pushover analysis or nonlinear time history analysis is performed to havea better estimate of demand on the structure due to a seismic event. Due to minimum flexural andshear design requirements, non-seismically designed structures may possess an inherent lateral strength
to resist minor to moderate earthquakes, but the performance of such structures under severe
earthquakes may be extremely poor (Bracci et al. 1995). A comparison of seismic design provisions
to the non-seismic design provisions points out to various deficiencies in the non-seismically detailed
structures. Weak column strong beam configurations, minimal transverse reinforced in columns forshear and confinement, especially in potential hinge zones, providing lap splices in potential hingezones, no ties in the joint core, and inadequate anchorage to beam bars in the joints are the mostcommonly found deficiencies. In this test program, the deficiencies that were included in the
structure comprised of no special confining reinforcement and no ties in the joint. In the third stage,
experiments were conducted on the structure with earthquake ground motions in two axes with a
gradually increasing level of PGA that represented earthquakes from low to moderate to severe
intensities.

2. Objectives and research significanceNon-conforming RC structures make the majority of RC structures existing worldwide, especiallyin developing countries, where the seismic detailing provisions were enforced only in 90’s. In such
structures, several nonlinearities exist apart from beam hinging and column hinging such as joint
shear failure, bond failure etc that make the modeling difficult. Also interactions between infill
panels and frame make things complicated. Shake table tests offer a way to have the valuable
insight to the structure in terms of the failure modes, study the interactions that can be utilized todevelop and validate the analytical models. Moreover such tests on realistic structures and groundmotions are the best way to prove the efficacy of the retrofit system. In view of the above the test
program was designed with following three primary objectives:
1. To investigate the usefulness of TLD to reduce the seismic response of realistic RC structures
under realistic ground motions
2. To investigate the seismic behavior and failure modes of non-conforming RC structure till failure.
3. To generate data for calibrating the available analytical models for better prediction of responseof such structures during earthquakes.
4. To study the influence of masonry infill panels on the stiffness and fundamental frequency of thestructure.

3. Experimental program
In this work, a 3D RC frame structure was tested under seismic ground motions. The structure
was symmetric in plan with 2 bays in both directions but in one direction, the structure was softer
than in other direction, due to the use of rectangular columns. The structure had a plinth storey,
whose height was 1/3rd of the height of other three stories. The test was conducted at the

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Akanshu Sharma, G.R. Reddy and K.K. Vaze

Fig. 2 Details of the test structure

Earthquake Engineering and Vibration Research Centre (EVRC) of Central Power Research Institute(CPRI) Bangalore. The shake table size is 3 m × 3 m and the payload capacity is 10 tons. Due tothe shake table size and weight restrictions, the structure had to be scaled down to approximately 1/
3rd scale. The section sizes of beams and columns were 75 mm × 100 mm with a 50 mm thick slab.
Fig. 2 shows the details and photographic view of the structure before the test. All the reinforcementof 6 mm or less was Mild steel reinforcement and 8 mm diameter bars were tor steel reinforcement.
The slab reinforcement consisted of 6 mm diameter bars at the rate of 100 mm c/c. The tested
average cube compressive strength of concrete was obtained as 33.0 N/mm2.
The structure originally had masonry infill completely filling the frame panels in X-direction,
whereas the frame panels in Y-direction (weaker direction) were not having any masonry infill. Thiswas mainly due to the weight restrictions of the shake table. The first set of the test was therefore
conducted on the structure with infill panels. Sine sweep tests were conducted on the structure witha PGA of 0.075 g, 0.1 g and 0.125 g in X, Y and Z (vertical) directions from 1 Hz to 50 Hz at therate of 1 Octave per minute, i.e. the frequency of sine wave was doubled in a minute. Fig. 3 showsShake table tests on a non-seismically detailed RC frame structure5

a typical input time history for the sine sweep tests corresponding to PGA of 0.1 g.
The masonry walls were then removed and almost equivalent masses were applied on the floors
and the sine sweep tests were again carried out. The criterion for placing the masses was such thatthe frequency of the structure could be brought to the desirable range of tuning with TLD. In thisway the fundamental frequency of the structure without and with walls was compared and thetuning between the sloshing frequency of TLD and fundamental frequency of the structure could be

Fig. 3 Typical input time history for sine sweep tests (PGA = 0.1 g)

Fig. 4 Structure with masses and TLD but without walls

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Akanshu Sharma, G.R. Reddy and K.K. Vaze

obtained. The second stage of tests consisted of tests on the structure without and with TLD kept on
the roof of the structure. The TLD were proportioned beforehand to have dimensions that would
make the sloshing frequency approximately equal to the estimated fundamental frequency of thestructure. Fig. 4 shows the photograph of the structure with removed walls, added masses and TLD
in the form of liquid containers at the roof. The added masses were 500 kg on the roof and 1st floorslab and 1000 kg on the 2nd floor slab. First, a series of tests were performed with the TLD andlater the TLD were removed and the same series of tests were performed without TLD to compare
the seismic response of the structure with and without dampers.
In the 3rd stage of the experimental program, biaxial ground motions were provided to the shake
table in X and Y directions simultaneously with equal excitations, while no input motion were givenin the Z direction. The time history was artificially generated such that the response spectrum
generated corresponding to the time history would closely envelope the target response spectrum(Fig. 5).

Fig. 5 Comparison of target and test response spectrum normalized to 0.1 g

Fig. 6 Accelerometer locations on the structure

Shake table tests on a non-seismically detailed RC frame structure

7

The tests started with the biaxial time histories provided to the table with a PGA of 0.1 g andgradually it was increased in steps of 0.1 g till 0.7 g. The limit of 0.7 g was due to the shake tablestroke limitations. The maximum available displacement stroke for the actuators of shake table was+/- 150 mm. The maximum displacement in either direction for a PGA of 0.1 g was 20 mm andTable 1 Test Matrix of experimental program
S. No.

Ground Motion

Axis of loading

PGA (g)Description

1

Sine sweep

X-Axis0.075

With infill panels

2

Sine sweep

Y-Axis0.075

With infill panels

3

Sine sweep

X-Axis0.1

With infill panels

4

Sine sweep

Y-Axis0.1

With infill panels

5

Sine sweep

Z-Axis0.1

With infill panels

6

Sine sweep

X-Axis0.125

With infill panels

7

Sine sweep

Y-Axis0.125

With infill panels

8Sine sweep

X-Axis0.1

Without infill panels, with added mass

9

Sine sweep

Y-Axis0.1

Without infill panels, with added mass

10

Simulated Seismic

X-Axis0.075

With TLD

11

Simulated Seismic

X-Axis0.1

With TLD

12

Simulated Seismic

X-Axis0.15

With TLD

13

Simulated Seismic

Y-Axis0.075

With TLD

14

Simulated Seismic

Y-Axis0.1

With TLD

15

Simulated Seismic

Y-Axis0.15

With TLD

16

Simulated Seismic

X-Axis0.075

Without TLD

17

Simulated Seismic

X-Axis0.1

Without TLD

18

Simulated Seismic

X-Axis0.15

Without TLD

19

Simulated Seismic

Y-Axis0.075

Without TLD

20

Simulated Seismic

Y-Axis0.1

Without TLD

21

Simulated Seismic

Y-Axis0.15

Without TLD

22

Simulated Seismic

X-Y

0.1

Biaxial ground motion

23

Simulated Seismic

X-Y

0.2

Biaxial ground motion

24

Simulated Seismic

X-Y

0.3

Biaxial ground motion

25

Simulated Seismic

X-Y

0.4

Biaxial ground motion

26

Simulated Seismic

X-Y

0.5

Biaxial ground motion

27

Simulated Seismic

X-Y

0.6

Biaxial ground motion

28

Simulated Seismic

X-Y

0.7

Biaxial ground motion

29

Sine sweep

Y-Axis0.1

To verify the change in frequency after
the induced damage in the structure

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Akanshu Sharma, G.R. Reddy and K.K. Vaze

correspondingly, the maximum displacement in either direction for a PGA of 0.7 g was around 140
mm, which was very close to the limit of the table and the tests could not be continued further. In
the end, a sine sweep test was conducted to study the frequency change of the structure due toinduced damage. Table 1 shows the test matrix of the experimental program.
In order to obtain the seismic response, accelerometers were mounted on the structure at various
locations as shown in Fig. 6 and were connected to the high speed data acquisition system. The
output readings from the accelerometers were recorded digitally at a frequency of 256 samples per
second.

4. Proportioning of TLD
The TLD are characterized by three major ratios (Chaiseri et al. 1989, Sun et al. 1991, Koh et al.
1994, Reed et al. 1998, Yu et al. 1999, Fujino et al. 1992, Banerji et al. 2000) namely tuning ratio,mass ratio and depth ratios. A TLD that reduces the structural response significantly for the given
set of values of these ratios may be considered as properly designed.
4.1 Tuning ratio
The fundamental linear sloshing frequency of the TLD is given by (Jin et al. 2007)
f = [{tanh(3.68∆)*3.68g/D}0.5]/2π(1)

Where, ∆ is the water depth ratio defined as ratio of undisturbed water depth h to the tankdiameter, D.
As the name suggests, tuning ratio of a TLD, is the ratio of the fundamental, linear sloshingfrequency, f, given by Eq. (1), to the natural vibration frequency of the structure. Generally a tuning
ratio of unity or very close to unity is considered to be optimum (Chaiseri et al. 1989, Sun et al.
1991, Reed et al 1998, Yu et al. 1999, Fujino et al. 1992, Banerji et al. 2000). Banerji et al.
concluded that it is reasonable to consider this tuning ratio to have a value of unity for strong
earthquake motions also. Based on these arguments, the target tuning ratio for the TLD was set tounity.
4.2 Depth ratio
As seen in Eq. (1), the only two parameters that can be varied to achieve the desired tuning ratio
are undisturbed water height, h and the tank diameter D. Thus, the water depth ratio ∆, which is theratio of water depth h to tank length D, is the significant parameter to mark the effectiveness of
TLD. Most of the earlier studies (Chaiseri et al. 1989, Koh et al. 1994, Yu et al. 1999) haverestricted this value to less than 0.1. However, Banerji et al. (2000) pointed out that the water
depths required for the short period structure with frequencies more than 2 Hz, and a tuning ratio ofunity are too small for practical implementation if such a restriction is maintained. They tested adepth ratio of 0.3 and concluded that the optimum value for such structures under high excitationlevels is around 0.3. They supported it with the argument that at a high excitation level, the energy
transmitted is large and even a TLD with relatively larger water depths dissipates energy throughsloshing and wave breaking. Therefore for strong earthquake motions, it is more effective to use

Shake table tests on a non-seismically detailed RC frame structure

9larger water depths, within the constraints of shallow water theory (Banerji et al. 2000). Based onthese arguments, the water depth ratio in this case was considered as 0.3.
4.3 Mass ratio
The mass ratio, is the ratio of the mass of water in the containers to the mass of the structure
itself. In order to make sure that TLD do not influence the dynamic characteristics of the structure
and also due to the practical limitations of space on the roof of the structure to place dampers, amass ratio of 1% was considered in this case.
4.4 Design of TLD
From pretest analysis, the fundamental frequency of the structure was estimated to be around 3.05 Hz
corresponding to design material properties. Since it was decided to consider tuning ratio as unity,
From Eq. (1), for a water depth ratio of 0.3 and f = 3.05, the diameter of the tank comes to about
80 mm. Therefore in this test, plastic containers with internal diameter of 80 mm were used with awater height of 24 mm. The total mass ratio in this case came out to a little less than 1% but due tospace constraints, it was considered suitable.

5. Test structure and prototype structure
In past, efforts have been put by researchers to perform such tests. Tests reported by El-Attar etal. (1991), Bracci et al. (1995), Lee and Woo (2002), Cardone et al. (2004), Dolce et al. (2005),
Dolce et al. (2006a, b, c), Hwang et al. (2006), Hashemi and Mosalam (2006a, b, 2007), Kakaletsis
et al. (2011) are a few examples of similar tests and many more tests may be found in literature.
However, one of the biggest technical limitations of the shake table tests is to design the structure
suitable enough to represent the real life structures without much distortion. Mostly this limitation
comes due to the size and weight capacity of the shake table. Though theoretically it would bepossible to design a specimen that can almost accurately simulate the real life structure at a smallerscale by following dimensional analysis (Buckingham 1914), the construction of a ‘true replica’model that satisfies all the similitude requirements needed by dimensional analysis is almost an
impossible task due to material limitations (Morcarz et al. 1981, Quintana-Gallo et al. 2010).
Therefore, the challenge is to design a least distorted model within the constraints of shake tablecapacity and material availability.
The structure tested in this experimental program represented a scaled down model of a part of athree storey reinforced concrete frame structure designed as per non-seismic design guidelines as
per Indian Standards (IS 456:2000). Although the structure was detailed non-seismically, it did not
represent a typical 70’s structure where plain round bars with hooks were used; rather it represented
structures built in later decades but not following seismic design and detailing provisions. Suchstructures make a large number of existing structures in India.
The design concrete cube strength was considered as 25 N/mm2 (corresponding to M25 specified
by IS 456:2000) and deformed bars having characteristic yield strength of 415 N/mm2 were selected
for the design. The choice of the set of materials is one of the most popular material parametersconsidered for design of such structures in India. The prototype structure was considered to have a10

Akanshu Sharma, G.R. Reddy and K.K. Vaze

Fig. 7 Structural layout of prototype structure considered for study

storey height of 3.0 meters and beam width of 4.0 meters, which is typical for a residential buildingin India giving a popular room size. The design was performed on the basis of the guidelines
provided by the code (IS 456:2000) with no seismic considerations. Fig. 7 shows the beam andcolumn layout of the hypothetical prototype structure considered for this study. The portion of thestructure enclosed within the dashed lines was considered to be scaled and tested on the shake table.
Although, ideally a scaled model of the prototype designed as per scaling laws and dimensional
analysis would be best suited, but as mentioned earlier due to the constraints set by shake tablecapacities and material availability, such a scaling is not possible. In this case, in general, a scale
factor of 3.33 was considered to design the model. The scale factor was basically governed by thesize limitation of the shake table. The table used for this experiment is 3 m × 3 m in size and therefore
the plan dimension of the structure to the outer of the columns was set as 2.5 meters, leaving justsufficient space for overhang of footings beyond columns. Thus, the center to center distance between
two columns was set as 1.2 meters leading to a linear scale factor, SL of 4/1.2 = 3.33.
6. Similitude requirements
6.1 Linear dimension scaling
As mentioned earlier, due to the shake table size limitation, a linear scale factor of SL = 3.33 wasenforced. Thus, the bay width was scaled from 4.0 meters to 1.2 meters and the storey height wasscaled from 3.0 meters to 0.9 meters. The plinth storey height was reduced from 1.0 meters to 0.3
meters. The beam and column dimensions for the prototype structure were maintained at 250 mm ×
325 mm so that they could be easily scaled to 75 mm × 100 mm.

Shake table tests on a non-seismically detailed RC frame structure

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6.2 Material scaling
As mentioned earlier, the prototype structure was considered to be built by using concrete of grade
M25 (characteristic cube strength = 25 MPa) and deformed bars with characteristic yield strength of
415 MPa. To avoid too much scaling from material point of view, it was decided to use concrete
mix of same grade. However, to maintain similitude, the maximum size of aggregate (msa) of 6 mm
was used (corresponding to msa of 20 mm in prototype). The design of the concrete mix yielded ina ratio of 1:1.35:2.02 with a water-cement ratio of 0.45. The mean concrete cube strength for the mix
was found to be 33.0 N/mm2, which was very close to the target mean strength of 33.75 N/mm2 asrecommended by IS Code (IS 456:2000). Thus, for concrete it can be said that the material scale
factor given by the ratio of concrete strength in the prototype to concrete strength in model, SFC wasunity.
In case of reinforcement, it was not possible to use deformed bars with characteristic yieldstrength of 415 MPa except as longitudinal bars in columns due to size availability issues. Due toarea scaling, the sizes for rebars other than longitudinal bars in columns were 6mm diameter or lessand only mild steel bars with characteristic yield strength of 250 MPa was available for these sizes.
Defining material scale factor for steel, SFS as the ratio of specified yield strength for rebars in
prototype to that in model, two different values of SFS were used. Thus, SFS for column main rebars
was unity and for other rebars, SFS = 415/250 = 1.66.
6.3 Reinforcement area scaling
According to similitude requirements (Sabnis et al. 1983), the required model reinforcement area
to provide the scaled bar yielding force is calculated asAp-2
Am = ------------------(2)SFS × SLThe model reinforcement area were scaled according to the law given by Eq. (2), though, aperfect scaling could not be done due to size and type of reinforcements available.
6.4 Mass scaling
Since the density and modulus of elasticity of the concrete in prototype and model were essentiallyequal, a mass simulation approach suggested by Quintana-Gallo et al. (2010) that corresponds to an
extension of what is suggested by Morcarz et al. (1981) was followed. The additional inertial masson the ith floor level in the model structure, ∆Mi corresponding to a unit mass in prototype is given
by Eq. (3).
1
1
∆Mi = -------2 ⎛ 1 – -----⎞
(3)
⎝
⎠
S
L
SL
Therefore for every ton of mass located at a storey level of the prototype structure, 0.063 tons ofmass was to be applied at the corresponding storey level for prototype structure. However, thisrestriction was not cent percent satisfied due to shake table weight restrictions. Moreover, as
mentioned in part 1 of the paper, another criterion for added mass was to bring the frequency of thestructure in the desirable range to test tuned liquid dampers.

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Akanshu Sharma, G.R. Reddy and K.K. Vaze

6.5 Comment on similitude
As shown above, similitude between the prototype and model was attempted for, following thescaling rules. However, it was found that every aspect of the structure could not be scaled due toshake table capacity and material availability constraints. But it should not be considered as a matterof big concern since the objective of the test was not to qualify the prototype structure by testing
but to study the behaviour of reinforced concrete structures, in general, under earthquakes.

7. Results and discussion
7.1 Results of 1st stage of experiments
7.1.1 Sine sweep tests on structure in plane of walls (X- direction)
The tests were carried out for a frequency range of 1 to 50 Hz, with acceleration levels of 0.075 g,
0.1 g and 0.125 g. During these tests, cracks in the mortar at the corner of walls were observed firston ground storey walls (at an acceleration of 0.1 g) and then on middle storey walls at an accelerationof 0.125 g). This is as expected since the maximum diagonal tension will come at the bottom
storey. Consequently, there was a gradual reduction in frequency for three different accelerationlevels as 9.75 Hz, 9.25 Hz and 8.75 Hz for 0.075 g, 0.1 g and 0.125 g respectively. Moreover, due
to degradation in the load carrying capacity and stiffness of the walls, the amplification from thebase to roof acceleration increased as 4.04, 4.5 and 5.2 for 0.075 g, 0.1 g and 0.125 g respectively.
However, no resonant frequency corresponding to second mode was obtained in any case. The
results of sine sweep tests in X-direction (in plane of the walls) are summarized in Table 2.
7.1.2 Sine sweep tests on structure in vertical direction
A single test with an acceleration level of 0.1 g was performed in Z-direction and no modes wereobserved in the range of 1-50 Hz.
7.1.3 Sine sweep tests on structure out-of-plane of walls (Y-direction)
The tests were carried out for a frequency range of 1 to 50 Hz, with acceleration levels of 0.075 g,
0.1 g and 0.125 g. During these tests, corner mortar cracks formed earlier due to in-plane tests forcorresponding accelerations propagated throughout the periphery of the walls first on middle storey
walls (at an acceleration of 0.1 g) and then on bottom storey walls at an acceleration of 0.125 g).
However, there was no significant cracking in the top storey walls. This clearly highlights the factthat the out-of-plane strength of the walls is significantly influenced by simultaneous action of inTable 2 Summary of results obtained during sine sweep tests in X-direction with walls
PGA

Visual Observation

Resonant Frequencyst

1 Mode2 ModeAmplificationnd

0.075 g

No cracking, No damage

9.75 Hz

Not Found

4.04

0.1

Cracks in mortar at corner of ground storey walls.

9.25 Hz

Not Found

4.5

0.125 gCracks in mortar at corner of middle storey walls.

8.75 Hz

Not Found

5.2

Shake table tests on a non-seismically detailed RC frame structure

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Table 3 Summary of results obtained during sine sweep tests in Y-direction with walls
PGA

Visual Observation

Resonant Frequencyst

1 Mode2 ModeAmplificationnd

0.075 g

No cracking, No damage

4.25 Hz17.0 Hz

13.69

0.1

Peripheral cracks in middle storey

4.25 Hz16.25 Hz

13.34

0.125 g

Peripheral cracks in bottom storey

4.00 Hz15.375 Hz

10.85

plane vibrations (as in the case of real earthquake scenarios). No significant change in the frequency
for first mode of the structure was noticed with the propagation of cracks. However, a reduction in
the second mode frequencies was observed with increase in acceleration levels. The frequencies ofthe structure were obtained as 4.25 Hz, 4.25 Hz and 4.0 Hz for first mode and 17.0 Hz, 16.25 Hz
and 15.375 Hz for the second mode against the acceleration levels of 0.075 g, 0.1 g and 0.125 g
respectively. The amplifications were 13.69, 13.34 and 10.85 against the acceleration levels of 0.075 g,
0.1 g and 0.125 g respectively. The results of sine sweep tests out-of-plane of the walls are summarizedin Table 3. It is interesting to note that amplifications were much higher in case of Y-direction,
where there are no walls. This is attributed to the fact that due to the presence of the masonry infill
panels, the behaviour of the frame structure is essentially changed from a moment-resisting frame
action to truss action (Murty and Jain 2000).
7.1.4 Sine-sweep tests on bare frame structure in X- and Y- directions
Sine sweep tests were conducted in two orthogonal horizontal directions in order to determine thefrequency of the structure after removal of walls and with added mass. It was found that thefrequency of the structure in X-direction has lowered to 3.5 Hz, and that in Y-direction is lowered to
3.0625 Hz.
7.2 Discussion on 1st stage of experiments
The 1st stage experiments clearly displayed that the masonry infill panels contribute greatly to thestiffness of the structure leading to higher natural frequencies. The increase of stiffness is generally
associated with attracting higher seismic forces for the cases of most of the real life reinforced
concrete framed structures. However, there is another important aspect to the contribution of masonrywalls, which is associated with the change of load transfer mode from predominant moment resisting
frame action for bare frames to predominant truss action for frames with infill walls. Therefore dueto the inclusion of walls, the moments at the ends of beams and columns are reduced while the axial
forces are increased.
Another important issue is the stability of walls itself. Masonry infill panels are traditionallyconstructed after the bare frame structure is ready, thereby not allowing any good bondage between
the walls and the frame members and the only thing resisting the forces at the interface is the weak
and thin mortar layer. As seen by the experiments, walls tend to develop brick-mortar interface
cracks at the corners at a relatively low PGA level under in-plane loads. These cracks propagate
through the periphery of the walls under out of plane loads making the walls unstable. Therefore,
under the combined action of in-plane and out-of-plane accelerations as would be the case in a realearthquake, the stability of the masonry walls are often endangered, which is one of the major

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Akanshu Sharma, G.R. Reddy and K.K. Vaze

causes for the loss of life and property during earthquakes. In such cases, even though the structural
framework may be stable but the walls itself may become unstable and lead to partial, nonstructural but life endangering failures. One solution to this problem is using dowels to anchor walls
to the frame or to prevent out of plane movement of the walls by using fibre mats etc. However,this is out of scope of this paper.
8. Results of 2nd stage of experiments
In the 2nd stage of the experiments, simulated seismic ground motion was applied to the shake
table. The accelerations were applied only in one direction at one time to the table, i.e., the groundmotion was essentially uniaxial. The input ground motion was the artificially generated time history,
response spectrum of which enveloped the desired response spectrum (Fig. 5). Fig. 8 shows thecorresponding acceleration time history actually provided to the shake table.
8.1 Simulated seismic tests on structure in X-direction with and without dampers
The simulated seismic tests were conducted on the structure in X-direction with and without
dampers for PGA levels of 0.075 g, 0.1 g and 0.15 g. The PGA level was limited to 0.15 g in order
to not induce any damage to the structure. Fig. 9 shows the typical acceleration response recorded at
the roof of the structure for the ground motion corresponding to 0.1 g PGA in X-direction without
dampers and Fig. 10 shows the same for the case with dampers. The maximum responses of thestructure were observed as 5.5 m/s2, 6.5 m/s2 and 8.7 m/s2 for the case of without dampers and as
4.2 m/s2, 5.1 m/s2 and 6.8 m/s2 for the case with dampers against the acceleration levels of 0.075 g,
0.1 g and 0.15 g respectively displaying that the TLD are capable to control the seismic response of
the structure.
8.2 Simulated seismic tests on structure in Y-direction with and without dampers
The simulated seismic tests were also conducted on the structure in Y-direction with and without

Fig. 8 Acceleration time history provided to the shake tableShake table tests on a non-seismically detailed RC frame structure

15

Fig. 9 Acceleration response of the structure without dampers recorded at roof in X-direction due to 0.1 g
PGA ground motion

Fig. 10 Acceleration response of the structure with dampers recorded at roof in X-direction due to 0.1 g PGA
ground motion

dampers for PGA levels of 0.075 g, 0.1 g and 0.15 g. The maximum responses of the structure wereobserved as 5.2 m/s2, 6.7 m/s2 and 8.5 m/s2 for the case without dampers and as 3.0 m/s2, 4.2 m/s2
and 5.8 m/s2 for the case with dampers against the acceleration levels of 0.075 g, 0.1 g and 0.15 g
respectively again displaying the suitability of TLD in controlling the response of the structure. The
summary of results of simulated seismic tests for X-direction is given in Fig. 11 and that for Ydirection is shown in Fig. 12.
The 2nd stage experiments displayed the efficiency of using tuned liquid dampers in controllingthe seismic response of reinforced concrete frame structures. As mentioned earlier, the response iscontrolled by tuning and closer the tuning ratio is to unity, in general, the response reduction should
be higher. The same effect is displayed in this case also. As discussed in the section of design ofTLD, the TLD were designed to have a sloshing frequency of around 3.05 Hz. From the sine sweep
tests performed in the two horizontal directions after removal of walls and added mass, it was foundthat the frequency of the structure in X-direction was 3.5 Hz, and that in Y-direction was 3.0625 Hz.
Therefore, of course the TLD were better tuned to the Y-direction frequency rather than the X-

16

Akanshu Sharma, G.R. Reddy and K.K. Vaze

Fig. 11 Maximum response of the structure for Xdirection without and with TLD

Fig. 12 Maximum response of the structure for Ydirection without and with TLD

direction frequency. Therefore, it can be expected that the response control is better in Y-direction
than in X-direction. Results summarized in Figs. 10 and 11 confirm this prediction. The average
reduction in the response was found to be around 22% in X-direction and around 37% in Ydirection. However, it must be remembered that in this case the excitation was given only in one
direction. If the structure would be excited in bidirectional motion, the response reduction even for
Y-direction would be lesser. Moreover in such cases when circular TLD are used, there is a risk forthe phenomena called ‘whirling’ to occur, which would result in reducing the effectiveness of TLD.
9. Results of 3rd stage of experiments
The third stage of tests started with simulated bidirectional earthquake with a PGA of 0.1 g. The
acceleration time history for 0.1 g PGA was shown in Fig. 8 and the corresponding displacement
time history for the ground motion is shown in Fig. 13. Same time history was provided in both X-

Fig. 13 Displacement time history provided to the shake tableShake table tests on a non-seismically detailed RC frame structure

17

Fig. 14 Floor acceleration time history for X-direction recorded at the roof level for a PGA of 0.1g PGA bidirectional base excitation

Fig. 15 Floor acceleration time history for Y-direction recorded at the roof level for a PGA of 0.1g PGA bidirectional base excitation

and Y- directions. The response of the structure recorded at roof in X-direction is shown in Fig. 14
and that in Y-direction is shown in Fig. 15. Similar time histories were recorded at each floor levelfor both directions corresponding to each PGA level. A plot of the maximum acceleration recordedv/s the PGA of base excitation for X- direction is shown in Fig. 16 and the same corresponding to
Y- direction is shown in Fig. 17.
The plots of Figs. 16 and 17 clearly show that at lower PGA base excitations, the maximum
response acceleration increase steeply with the base excitation suggesting a linear behaviour of thestructure, whereas as the base excitation is increased, the response becomes flat suggesting inelastic
behaviour.
It can be observed from Figs. 16 and 17 that both the curves display an almost linear increase of
maximum recorded roof acceleration in both the directions till a PGA level of 0.4 g. However, as
can be noticed, the acceleration values are not directly proportional to the PGA values, e.g., the18

Akanshu Sharma, G.R. Reddy and K.K. Vaze

Fig. 16 Recorded peak roof acceleration v/s PGA for
X-direction

Fig. 17 Recorded peak roof acceleration v/s PGA for
Y-direction

Fig. 18 Crack patterns observed during initial stages of the test

recorded maximum acceleration does not get doubled when the PGA is doubled. This essentiallysuggests, though the behaviour is linear, it is not elastic, i.e., certain damage must have occurred in
the system, which would have changed the dynamic characteristics of the structure thereby reducing
the response, but the damage was not so severe to cause any nonlinearity in the global response of
the structure. This can be further confirmed with the pictures shown in Fig. 18.
Fig. 18 (a) shows the very first cracks that appeared at the base of the columns, close to thefooting. This definitely is as expected since due to the lateral loads generated by the ground motion,
maximum moments are experienced at the base of the columns and therefore the first cracks appear
there. However, as seen in Fig. 18 (a), the cracks are only initiated and they only indicate that theconcrete tensile stresses are exceeding their limits but definitely there is no yielding of the reinforcement.
Therefore, due to the cracks, it is expected that the stiffness is reduced but the structure shall remain
essentially linear. This explains a linear but not proportional rise in the maximum response of thestructure corresponding to increasing PGA.
As the PGA level of the base excitation was increased to 0.3 g and 0.4 g, a few more cracks started
to appear. For a PGA of 0.3 g, cracks appeared on the beams of first floor at the beam-column interface(Fig. 18 (b)). However, again the cracks did not suggest any yielding of bars. At 0.4 g, the beam

Shake table tests on a non-seismically detailed RC frame structure

19

Fig. 19 Damage patterns observed at PGA = 0.5 g

flexural cracks opened up significantly and some spalling was also observed (Fig. 18 (c)) suggesting
the initiation of yielding of rebars inside. Since the cracks suggest a yielding of bars, linear responseof the structure can no longer be expected. Therefore, now the dynamic response of the structure
must be modified not only because of the change in dynamic characteristics of the structure but also
due to the additional hysteretic damping coming into picture, that should further reduce the responseof the structure. This is exactly what can be observed from Fig. 16 and 17 that after the PGA of 0.4 g,
the response curves for the structure tend to become flatter.
Fig. 19 shows the damage patterns observed for the base excitation with a PGA level of 0.5 g. It
was observed that many new cracks were formed in the beams and columns and the old cracks
opened widely and led to spalling of concrete. The major damages were observed for the columnsof 1st and 2nd storey levels. This definitely suggests that the reinforcement in these critical locations(such as end of beams and columns) have yielded. However, there was a very interesting thingobserved that the cracks formed at the very base of the columns at the PGA of 0.2 g (Fig. 18 (a))
neither grew any further nor opened significantly. This seems to be something unexpected in thefirst look, since these areas seem to be the most critical and generally the first hinges would be
expected to form at the base of the columns. It was observed that instead of the cracks developed at
the column bases opening up, the newer cracks that were formed at the plinth beam level (Fig. 19)
became significant. This clears some of the doubts, since the presence of the plinth beam shifts thecritical zones from the base of the columns to the plinth level. This is because the presence of plinth
beam reduces the shear span for the columns and therefore instead of having a flexural mechanism,
the column tends to develop a strut type of mechanism, thereby avoiding any further damage at thecolumn base. This suggests that plinth beam definitely has beneficial effects as it prevents the
structure to form a soft story type of mechanism. Such a mechanism might be possible in this case
had the plinth beam not been there because in such a case, after the hinge formation at columnbases, general tendency of next hinge formation is at the first story column level that may finallylead to a storey mechanism. However, these failures at the end of columns and beams, which aremainly flexural type, definitely indicate significant energy dissipation through hysteresis leading to
reduced response.
After the PGA of 0.5 g, the spalling was significant and the already formed cracks were seen to
open up widely. Fig. 20 shows the damage patterns observed at the PGA of 0.6 g and Fig. 21
shows that for the PGA of 0.7 g, which was the maximum base acceleration level applied. In thefirst photograph of Fig. 20, the bottom beam bars are visible and in the third photograph it can be20

Akanshu Sharma, G.R. Reddy and K.K. Vaze

Fig. 20 Damage patterns observed at PGA = 0.6 g

Fig. 21 Damage patterns observed at PGA = 0.7 g

seen that the interior column just below the first storey level has lost almost all of the concrete
forming a total hinge. Not only the cover concrete has spalled off but even the core concrete within
the reinforcement is lost. This shows the effect of lack of proper confinement with ties in thecritical zones. However, the failure is essentially a flexural failure that may be attributed to large
span to depth ratio for the columns, but the plastic hinge length is quite large in this case.
Fig. 21 mainly concentrates on showing the damages in the joints. The first photograph shows atypical shear failure of the interior joint. However, it must be noted that in this case it is not a purejoint shear (JS) failure but more of a so-called BJ failure mode where the members framing into thejoint yield first and the shear failure of the joint occurs later. Similarly joint shear cracks can beseen propagating in the exterior joint shown in the last photograph, which is the same joint that isshown in the first picture of Fig. 13. These pictures indicate that the joints could not remain
essentially elastic, which is mainly due to the lack of shear reinforcement in the joint core, but inthis case, the joint failure was not so critical since the members of the structures failed first andthen the cracks in the joints were induced.
As mentioned earlier, the tests could not be carried out for further higher PGA values due to thelimitation of the stroke of the shake table. However, as seen above, the structure already sufferedsignificant damage and therefore, not being able to carry out the tests further was not a big loss. As
shown in Table 1, after the completion of simulated seismic tests, a sine sweep test was conductedin Y-direction to obtain the frequency of the damaged structure. This test, which was carried out

Shake table tests on a non-seismically detailed RC frame structure

21

with a PGA of 0.1 g showed that the structural frequency in Y-direction has come down to 1.025
Hz. As mentioned earlier, the original fundamental frequency of the structure in Y-direction was
3.0625 Hz, which means the structural frequency is reduced to almost 1/3rd due to the damage.
Since the mass of the structure is not significantly varied (except for a small spalling), this reduction
in frequency points to a corresponding reduction in the stiffness of the structure by 9 times. This
reduction is stiffness is attributed partly to the reduced modulus of the materials and partly to thechange in load resisting mechanism due to hinge formations.

10. Conclusions
In this experimental program, a 3D RC frame structure was tested on the shake table to study theeffects of infill panels on the dynamic characteristics of the structure, to verify the efficiency of
TLD system to control the seismic response of the structure and to study the nonlinear behaviour of
RC structure under increasing dynamic loads till failure. The studies on the effect of masonry infill
further verified that the masonry infill panels contribute greatly to the stiffness of the structureleading to higher natural frequencies and change in load transfer mechanism from moment frame
action to truss action. However, the vulnerability of the walls themselves from stability point ofview was another important issue pointed out in the program. It was seen during the experimentsthat walls tend to develop interface cracks at the corners at a relatively low PGA level under inplane loads, which further propagate through the periphery of the walls under out of plane loads
making the walls quite unstable. The solution to this problem may be proper doweling of walls withthe frame. The studies on TLD suggested that a properly designed TLD can offer a very goodsolution for controlling the seismic response of the RC frame structures under seismic excitation.
The TLD is probably one of the least invasive solutions for retrofitting and can be installed byvirtually not disturbing the occupants of the building at all. However, as with all other retrofit
solutions, it has its own limitations too. One major limitation is to use TLD for relatively stiffer
structures. As seen in this case, the damper size for a structure with fundamental frequency of 3Hz
is only 80mm, which may seem to be too small for real life buildings. For softer structures, the size
increases as the square of time period and the sizes become more practical. However, for structures
having fundamental frequencies equal to or more than 2 Hz, a higher water depth ratio such as 0.3
may lead to better sizes. Also, under strong seismic excitations, such water depth ratios are found tohave better response control, even though they strictly do not fall under the shallow wave theory
applicability criteria.
In the third stage of tests under gradually increasing seismic loads, the responses were recorded atvarious levels due to the ground motions and the same were compared with the PGA of the baseexcitation. The curves suggested a gradual induction of damage in the structure, which was also
confirmed by physically observing the damage patterns. It was observed that the very first cracks
appear at the base of the columns but these cracks do not grow further, which may be attributed tothe presence of plinth beams. Instead most of the latter cracks are formed at the first storey level
which can be attributed to the high inertia at the storey level.
In the beginning the response of the structure raised steeply, though not proportionally, to the
PGA level, which is attributed to gradual decrease in the elastic modulus of the material due tocracking. Later, as the PGA level was increased, the response of the structure was found to increase
nonlinearly with respect to the PGA, which happened due to the higher damping achieved by the

22

Akanshu Sharma, G.R. Reddy and K.K. Vaze

structure due to hysteretic energy dissipation. Mostly the damage could be classified as flexural
failures of beams and columns, with some contribution of shear and torsion modes. Large spallingof concrete in the critical regions at the ends of beams and columns was observed, even from thecore, due to the lack of confinement by lateral ties. In the end, the beam-column joints of thestructure developed shear cracks displaying the so-called ‘BJ’ failure mode.
Although from the test results it seems that the model structure could resist an earthquake with a
PGA level of 0.7 g, this may be misleading. This is because, the base acceleration was gradually
increased in this case and at for acceleration level, certain damage occurred that altered the stiffnessof the structure and also enhanced the damping through hysteretic energy dissipation. If the structure
would have been exposed directly, in the as-built condition, to the higher PGA levels, the response
could have been quite different and the structure might actually withstand a lower acceleration level
only. However, as mentioned earlier, the objective of this test was not to qualify a real life structure
by testing but to observe the various failure modes and sequence of events that may take place insuch structures under earthquakes. Still, it may be concluded that the model structure could resist
the low to moderate earthquakes reasonably well, but the performance of the structure under severe
earthquakes was poor.

Acknowledgements
The experiments were carried out under the research project funded by BARC to CPRI. The
authors are highly thankful to the whole team at EVRC, CPRI that worked enthusiastically to
conduct the tests and generate the useful information. The authors acknowledge the support andguidance of Dr. A.K. Ghosh and Mr. H.S. Kushwaha to conduct the tests.

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