ACI STRUCTURAL JOURNALTECHNICAL PAPER

MS No. S-2013-022.R2

Design Implications of Large-Scale Shake-Table Test on
Four-Story Reinforced Concrete Building
by T. Nagae, W. M. Ghannoum, J. Kwon, K. Tahara, K. Fukuyama, T. Matsumori, H. Shiohara,
T. Kabeyasawa, S. Kono, M. Nishiyama, R. Sause, J. W. Wallace, and J. P. Moehle
A full-scale, four-story, reinforced concrete building designed inaccordance with the current Japanese seismic design code wastested under multi-directional shaking on the E-Defense shake
table. A two-bay moment frame system was adopted in the longer
plan direction and a pair of multi-story walls was incorporated
in the exterior frames in the shorter plan direction. Minor adjustments to the designs were made to bring the final structure closer to
U.S. practice and thereby benefit a broader audience. The resulting
details of the test building reflected most current U.S. seismic
design provisions. The structure remained stable throughout theseries of severe shaking tests, even though lateral story drift ratios
exceeded 0.04. The structure did, however, sustain severe damage
in the walls and beam-column joints. Beams and columns showed
limited damage and maintained core integrity throughout the seriesof tests. Implications of test results for the seismic design provisions of ACI 318-11 are discussed.
Keywords: collapse; damage; design; full-scale; moment frame; multistory; shake table; shear wall.

INTRODUCTION
Code requirements for reinforced concrete have evolved
significantly around the world in the past decades. In the
United States, the 1971 San Fernando, CA, earthquake wasa watershed event leading to the introduction of requirements for ductile reinforced concrete buildings, whichhave evolved incrementally since that time based on field
and laboratory experiences. In Japan, following a history ofseveral damaging earthquakes and many laboratory tests, theJapanese seismic design code was substantially revised in
1981. In the 1995 Hyogoken-Nanbu earthquake, many reinforced concrete buildings designed before 1981 experienced
major failures, especially in the first-story columns and walls.
Although newer reinforced concrete buildings designed inaccordance with the revised 1981 code showed improved
resistance against collapse, several sustained severe damage
due to their large deformations. Such damage made it difficult to continue using them after the earthquake and resulted
in high repair costs. This experience demonstrates thatfurther improvements in seismic design of concrete buildings might be desirable for the future.
It was in light of the aforementioned experiences that alarge-scale shake-table testing program was conducted in2010. Within the program, a full-scale, four-story, reinforced concrete building designed in accordance with thepresent Japanese seismic design code was tested by using
the E-Defense shake table. The main objectives of the study
related to the concrete building were: 1) to verify methods
for assessing performance such as strength, deformation
ACI Structural Journal

capacity, and failure mode; 2) to identify suitable computational methods to reproduce the seismic responses of thebuilding; and 3) to develop a practical method for assessing
damage states regarding reparability.
Design and instrumentation of the test structure were
performed with input from U.S. co-authors. Wherever
possible, minor adjustments to the designs were made tobring the final structure closer to U.S. practice and thereby
benefit a broader audience. The resulting details of the test
building reflected the most current U.S. seismic design
provisions (Nagae et al. 2011b).
Summaries of the global behavior of the test building
and key local damage and deformation observations arepresented. A comparison between the details of the test
structure and U.S. seismic design practices is also provided.
Implications of test results for the seismic design provisions of ASCE 7-10 ASCE/SEI Committee 7 2010) andACI 318-11 (ACI Committee 318 2011) are discussed. In a
related publication (Nagae et al. 2011a), the seismic design
provisions of the Architectural Institute of Japan (AIJ 1999)
were evaluated in light of test results.
RESEARCH SIGNIFICANCE
Current Japanese and U.S. seismic design provisions are
based on pseudo-dynamic component tests, sub-assembly
tests, and limited dynamic tests of partial structural systems.
The test presented is a first-of-its-kind, multi-directional,
dynamic test of a complete, full-scale reinforced concretebuilding system to near collapse damage states. The test
provides unique data on component and system performancethat are used to evaluate current seismic design provisions
and highlight potential code changes.
SPECIMEN DETAILS
Figure 1 shows the plans and framing elevations of thereinforced concrete test building. Figure 2 shows a photograph of the test building on the E-Defense shake table. The
height of each story is 3 m (118.1 in.). The building footprint
measures 14.4 m (47 ft 3 in.) in the longer (X) direction, and
7.2 m (23 ft 7.5 in.) in the shorter (Y) direction. A two-bay
moment frame system was adopted in the longer (X) plan
direction and a pair of multi-story walls were incorporated
ACI Structural Journal, V. 112, No. 1-6, January-December 2015.
MS No. S-2013-022.R2, doi: 10.14359/51687421, received May 21, 2014, andreviewed under Institute publication policies. Copyright © 2015, American Concrete
Institute. All rights reserved, including the making of copies unless permission is
obtained from the copyright proprietors. Pertinent discussion including author’s
closure, if any, will be published ten months from this journal’s date if the discussion
is received within four months of the paper’s print publication.

1

Fig. 1—Framing and reinforcing details. (Note: Dimensions are in mm; 1 mm = 0.039 in.)Fig. 2—Reinforced concrete (left) and prestressed concrete(right) specimens on the E-Defense shake table.
in the exterior frames in the shorter (Y) plan direction. The
thickness of the top slab was 130 mm (5.1 in.). Rigid steel
frames were set within the open stories of the test specimen
for collapse prevention and measurement of story deformations. Representative building mechanical equipment2

was incorporated to assess potential damage during strong
seismic motions. Table 1 lists the various weights of the test
specimen. The weight was estimated based on the reinforced
concrete members, the fixed steel frames, and the equipment. Figure 1 shows dimensions and reinforcement detailsof typical members. The test building was designed in accordance with current Japanese seismic design practice.
When constructing the test building, columns, walls,
beams, and the floor slab were cast monolithically. The
longitudinal reinforcement of columns, beams, and the wallboundaries were connected by gas pressure welding. Lap
splices were used for the reinforcement of other parts of
the walls and the floor slabs. The frames in the test building
were nominally identical in design and detailing. The shear
walls at axes A and C contained the same amount of longitudinal reinforcement but differed in the spacing of transverse
reinforcement (Fig. 1). A complete set of drawings and specimen details can be found in Nagae et al. (2011b). Additional
test data can be found on the NEEShub website (NEEShub2011) and in Tuna (2012).
SPECIMEN DESIGN
The extent to which the test structure satisfies the seismic
design provisions of ASCE 7-10 and ACI 318-11 is explored
ACI Structural Journal

Table 1—Weight and design forces(A) Structural elements, kN

RC

Roof

Fourthfloor

Table 2—Material properties of concrete
Third
floor

Second
floor

σB,
N/mm2

Ec,
N/mm2

Column

53

106

106

106

Cast of fourth story and roof floor slab

27

41.0

30.5

Beam

240

240

240

240

Cast of third story and fourth floor slab

27

30.2

30.3

Wall

40

79

79

79

Cast of second story and third floor slab

27

39.2

32.8

Slab

484

428

424

420

Cast of first story and second floor slab

27

39.6

32.9

Sum

816

853

849

845

Roof

Fourth
floor

Third
floor

Secondfloor

Stair and handrail

6

6

6

6

Measurement frame

0

3

17

17

Equipment

112

5

0

0

Sum

118

14

23

23

Total of (A) and (B), kN

934

867

872

867

Fourth
story

Third
story

Second
story

First
story

ΣWi, kN

934

1801

2673

3541

Ci = 0.2 × Ai

0.29

0.25

0.22

0.20

Qi, kN

273

450

593

708

(B) Non-structuralelements, kN
Steel

Fc,
N/mm2

Notes: 1 kN = 0.225 kip; Wi is weight of floor i; Ai is shape factor for vertical distribution of lateral forces for floor i; Ci is lateral force at floor i as a fraction of ƩWi; andQi is shear at story i.

in this section. The building specimen was designed to withstand the seismic lateral forces presented in Table 1 (MLIT
2007) without members exceeding their elastic limits. These
forces, which sum to 20% of the weight of the structure, arehigher than those that would be specified by ASCE 7-10
(Section 12.8.1.3), which caps seismic lateral forces fora low-rise building to 1/R times the structure weight for adesign basis earthquake, where R is the response modification coefficient (8 for special reinforced concrete moment
frames and 6 for special reinforced concrete shear walls).
The vertical distribution of the design forces, given by theparameter Ai in Table 1, is similar to the ASCE 7-10 specification (approximate inverted triangular distribution).
Results of material tests are given in Tables 2 and 3. In
subsequent evaluations, the moment and shear strengthsof each member were calculated adopting the compressive
strength of concrete and the yield strength of steel reinforcement obtained by averaging material test results.
To aid in the design of the test specimen, pushover(nonlinear static) analyses were conducted on line-element
models of the structure. Figure 3 presents pushover results
for the final test specimen details. The analytical model used
for pushover analyses was built following work by Kabeyasawa et al. (1984). The effective flange width of a top slab
was adopted in accordance with the recommendations of
the 2007 MLIT Standard. A vertical distribution defined by
the parameter Ai (Table 1) was adopted for the lateral force
distribution. In the analytical model, inelastic deformationsof beam elements were represented by rotational springsat the ends of elements. The first and second break points
corresponding to member cracking strength and flexuralstrength were assigned in the tri-linear moment-rotation
ACI Structural Journal

Notes: 1 N/mm2 = 0.145 ksi; Fc is specified concrete compressive strength; σB ismeasured concrete compressive strength; and Ec is measured secant modulus of concrete.

Table 3—Material properties of steel
D22

Grade

Anominal, mm2

σy, N/mm2

σt, N/mm2

Es, kN/mm2

SD345

387

370

555

209

D19

SD345

287

380

563

195

D13

SD295

127

372

522

199

D10

SD295

71

388

513

191

D10

SD295

71

448

545

188

D10

KSS785

71

952

1055

203

Notes: 1 mm2 = 0.0016 in.2; 1 N/mm2 = 0.145 ksi; Anominal is nominal area of reinforcing bars; σy is measured yield strength of steel reinforcement; σt is measured
ultimate strength of steel reinforcement; and Es is measured elastic modulus of steel
reinforcement.

relationship. The secant stiffness corresponding to the flexural strength was calculated in accordance with provisionsof the MLIT standard (2007). Beyond flexural yielding, thestiffness was reduced to 0.01 times the initial effective stiffness. The pushover analysis indicates that the ultimate baseshear strength of the building specimen is approximately
0.42W (1500 kN [337 kip]) in the frame direction and 0.51W(1800 kN [405 kip]) in the wall direction.
Figure 4 shows the column-beam moment strength ratios.
Reinforcement of the top slab was reflected in the moment
strength of beams in negative bending (top in tension). Effective flange widths of beams were adopted in accordance withthe recommendations of the 2007 MLIT Standard or ACI
318-11, which produced roughly similar flange widths. Variations of column axial forces due to lateral forces were estimated from pushover analysis in the Japanese calculations.
In the U.S. calculations, a plastic mechanism was assumed
in which hinging of the columns occurs at the foundationand just below the roof, and beam hinging occurs at column
faces at intermediate floors in the frame direction. In thewall direction, the assumed plastic mechanism consideredhinging of the columns and walls at the foundation, and beam
hinging at column and wall faces. Discrepancies in columnbeam moment strength ratios evaluated using ACI and MLIT
procedures (Fig. 4) can mostly be attributed to differences inthe estimates of axial forces on columns. From the second to
fourth floors, the column-beam moment strength ratios wereslightly below 1.0 for interior columns, while those of exterior columns ranged from approximately 1.0 to 1.87.
Assessment of specimen design in accordancewith U.S. seismic design practiceThe structure was assessed in both the X and Y directions using ACI 318-11 and ASCE 7-10 provisions. The
3

Fig. 3—Pushover analysis results. (Note: 1 kN = 0.225 kip.)Fig. 4—Moment strength ratios of columns to beams.
goal was to determine how well the structure compareswith U.S. seismic design practices. Rather than presume
that the building was to be constructed at a particular sitewith corresponding site seismic hazard, the assessments ofseismic design requirements are based on a seismic hazard
represented by the linear response spectrum for the 100%
JMA-Kobe ground motion to which the test structure wassubjected.
Shear wall direction (y-direction)—The approximate
natural period in the shear wall direction is 0.31 seconds
based on Eq. 12.8-7 in ASCE 7-10. The spectral acceleration
corresponding to this period is approximately 2.5g for the
100% JMA-Kobe ground motion imparted to the structure(Fig. 5, y-direction). Elastic analysis was performed using
equivalent (static) lateral forces corresponding to the spectral acceleration divided by an R factor of 6, as specified in
ASCE 7-10 for a building frame system with special rein4

forced concrete shear walls. Equivalent lateral forces were
distributed over the height of the structure in accordance with
provisions of ASCE 7-10. An effective moment of inertia
equal to 50% of the gross moment of inertia was used overthe full wall height: an intermediate value between the effective moments of inertia provided in ACI 318-11 for cracked
and uncracked walls. Selected wall effective moments ofinertia are also consistent with values recommended by
ASCE 41-06 (ASCE/SEI Committee 41 2007a) for cracked
walls. An effective moment of inertia equal to 30% of thegross moment of inertia was used for beams and columns
as per ASCE 41-06 (ASCE/SEI Committee 41 2007b) –supplement 1) provisions for beams and columns with low
axial loads. Beams were considered T-beams with an effective flange width evaluated in accordance with provisions of
ACI 318-11. Joints were taken as rigid. Elastic analysis of
the walls decoupled from frames at Axes A and C indicates
ACI Structural Journal

Fig. 5—Acceleration response spectra of input waves. (Note:
Damping ratio = 0.05; 1 m/s2 = 39.37 in./s2.)
that the walls would develop their design moment strength(0.9 × nominal moment strength) at approximately 0.37/Rof the JMA-Kobe 100% motion. If wall-frame interaction istaken into account, however, the wall-frame system would
develop its design moment strength at approximately 0.55/Rof the 100% JMA-Kobe motion. Thus, the building in thewall direction has only 55% of the strength that would be
required for the JMA-Kobe motion if that motion is considered as the design earthquake shaking level. In subsequentdiscussion, wall-frame interaction is taken into account.
When applying the equivalent lateral-force distribution inaccordance with ASCE 7-10, wall flexural yielding occurs
at a lower load than that generating the wall’s factored shear
strength. Distributed vertical and horizontal steel satisfied all
shear reinforcement requirements of ACI 318-11.
The wall-foundation interface was not intentionallyroughened prior to casting the walls. Given the amountof longitudinal steel crossing the interface, the axial force
on the walls, and a friction coefficient of 0.6, nominal
shear-friction strength in accordance with ACI 318-11 ofboth wall bases was approximately 2140 kN (482 kip). Thatshear-friction strength exceeded estimated shear demands byapproximately 55% based on the 100% JMA-Kobe groundmotion. Nominal shear-friction strength was, however, only
20% higher than maximum base shear demand estimatedfrom pushover analysis (approximately 1800 kN [405 kip]),
which accounts to some extent for member over-strength.
ACI 318-11 allows the use of two methods to determine
if boundary elements are required in walls. If the drift-based
method is considered (ACI 318-11, Section 21.9.6.2), noboundary elements are required in the walls for the 100%
JMA-Kobe motion, whether drift estimates are obtained
considering wall-frame interaction or not. If the stressbased method is considered (ACI 318-11, Section 21.9.6.3),
however, boundary elements are required in the walls up toa height of 7550 mm (297 in.) from the base of the wall ifwalls are considered decoupled from the frames, and a heightof 5060 mm (199 in.) if wall-frame interaction is accounted.
If one considers that boundary elements are not required in
the walls, minimum boundary detailing in both walls satisfies ACI 318-11 provisions. If one considers that boundary
elements are required, however, the provided spacingof hoops in the boundary elements of the wall at Axis C(100 mm [3.94 in.]) marginally exceeds the required spacing
ACI Structural Journal(83 mm [3.26 in.]). In the wall at Axis A, hoops were spaced
at 80 mm (3.15 in.) in the first story and this spacing satisfies all ACI 318 hoop spacing requirements for the boundary
element. In the upper stories of the wall at Axis A, hoops
in the boundary regions were spaced at 100 mm (3.93 in.)and therefore did not satisfy the ACI 318-required spacingof 83 mm (3.26 in.).
If wall-frame interaction was considered, beams spanning
between shear walls and corner columns were found to havesufficient moment strength to resist moments from elastic
analysis based on the 100% JMA-Kobe motion hazard level.
Shear strengths of the beams were sufficient to develop
beam probable moment strengths.
Because demands on corner columns in the shear wall
direction were significantly lower than demands on the same
columns in the frame direction, capacity and detailing ofcorner columns will be described in the section discussing
the frame direction (x-direction).
Frame direction (x-direction)—The approximate natural
period in the moment frame direction is 0.44 secomds basedon ASCE 7-10 Eq. 12.8-7. The spectral acceleration corresponding to this period is approximately 1.45g for the 100%
JMA-Kobe ground motion imparted to the structure (Fig. 5,
x-direction). Elastic analysis was performed using equivalent(static) lateral forces corresponding to the spectral acceleration divided by an R factor of 8, as specified in ASCE 7-10
for special reinforced concrete moment frames. Equivalentlateral forces were distributed over the height of the structurein accordance with ASCE 7-10. Elastic analysis of the frames
indicates that the first-story corner columns reach design
flexural strength at a shaking level corresponding to approximately 1.4/R of the JMA-Kobe 100% motion. All frame
member strengths therefore exceeded the required design
strength corresponding to a 100% JMA-Kobe hazard level.
Factored shear strengths of all beams were not sufficient
to develop probable moment strengths due to the requirement that concrete shear contribution be taken as zero
(ACI 318-11, Section 21.5.4.2). Maximum beam shear
stresses corresponding to the development of probable
moment strengths ranged from 2.0 to 2.7 times the squareroot of the concrete compressive strength in psi (0.17 to
0.22 MPa). The spacing of beam transverse reinforcement
was 200 mm (7.87 in.) in the critical plastic hinge regions,which exceeds the maximum allowable spacing of 120 mm(4.72 in.) as required by ACI 318-11.
Factored shear strengths of the third- and fourth-story
columns were not sufficient to develop probable moment
strengths. Column shear stresses corresponding to the development of column probable moment strengths ranged from
1.4 to 3.8 times the square root of the concrete compressive
strength in psi (0.114 to 0.315 MPa). Column-end transverse reinforcement met spacing and layout requirements of
ACI 318-11 in the first two stories but not the top two stories.
No columns met the requirement for minimum volumetricreinforcement ratio in the critical end regions; columns had
20 to 50% of the hoop volumes required by ACI 318-11
in the critical end regions. Transverse reinforcement ratios
varied substantially between columns in different stories due
to differences in numbers of crossties.
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Table 4—Key response values at roof
Test No.

Maximum roof acceleration

Maximum roof drift*

Residual roof drift

Input wave

x-direction, m/s

y-direction, m/s

x-direction, mm

y-direction, mm

x-direction, mm

y-direction, mm

JMA-Kobe 25%

3.12

6.37

16.9

24.2

0.5

0.4

2

JMA-Kobe 50%

7.03

11.01

122.4

106.9

1.1

5.4

3

JMA-Kobe 100%

9.65

14.01

242.7

323.9

6.2

22.5

4

JR-Takatori 40%

6.46

8.13

240.4

240.8

1.3

7.9

5

JR-Takatori 60%

8.09

9.99

278.1

414.0

8.0

11.6

1

2

2

*Maximum roof drifts do not include residual drifts accrued from previous tests.
Notes: 1 m2/s = 39.37 in./s2; 1 mm = 0.039 in.

satisfied the 6/5 minimum requirement of ACI 318-11. That
requirement was not satisfied at interior joints.
E-DEFENSE SHAKE-TABLE FACILITY AND TESTCONDITIONSThe E-Defense shake-table facility has been operated
by the National Research Institute for Earth Science andDisaster Prevention of Japan since 2005. The table is 20 x
15 m (65 ft 7 in. x 49 ft 3 in.) in plan dimension and can
produce a velocity of 2.0 m/s (78.7 in./s) and a displacement of 1.0 m (39.4 in.) in two horizontal directions simultaneously. It can accommodate a specimen weighing up to
1200 tonnes (1323 tons). In this series of tests, the considered reinforced concrete building was tested side-by-side
with a prestressed concrete building having almost the same
configuration and overall dimensions (Fig. 2). More detailabout the test structure, including detailed drawings, can befound in Nagae et al. (2011b).
Fig. 6—Maximum inter-story drift distribution.
Joint shear demands for both interior and exterior joints
were calculated considering force equilibrium on a horizontal plane at the midheight of the joints, in accordancewith ACI 318-11. Joint shear demands calculated including
the contribution of slab flexural tension reinforcement within
the ACI 318 effective flange width were found to be approximately 20 to 40% higher than demands computed ignoring
the slab contribution. Note that ACI 318 does not requireconsideration of the slab reinforcement in calculations ofjoint shear demand. Regardless of whether slab contribution
was taken into account, all joint design shear strengths, basedon ACI 318-11, exceeded joint shear demands. Because
joints were only confined by hoops without crossties, themaximum center-to-center horizontal spacing between hoopor crosstie legs was larger than the ACI 318-11 limit of
350 mm (14 in.). The provided hoop spacing in the jointsof 140 mm (5.5 in.) was larger than the maximum spacing
allowed by ACI 318-11 of approximately 25 mm (1 in.) forthe provided arrangement of hoops without crossties (limitedby minimum volumetric reinforcement ratio requirements).
Other joint detailing satisfied ACI 318-11 requirements,
including those for longitudinal bar anchorage.
Figure 4 shows column-beam nominal moment strength
ratios. Below the roof, all strength ratios for exterior columns

6

LOADING PROGRAM
Ground motions designated as JMA-Kobe and JR-Takatori, recorded in the 1995 Hyogoken-Nanbu earthquake,were adopted as the input base motions. The North-South-direction wave, East-West-direction wave, and vertical-direction wave were input to the y-direction, x-direction, andvertical direction of the specimen, respectively. The intensity of input motions was gradually increased to observedamage progression. The adopted amplitude scaling factors
for JMA-Kobe were 10, 25, 50, and 100%. Following the
JMA-Kobe motions, the JR-Takatori motion scaled to 40
and 60% was applied to impart large cyclic deformations.
Figure 5 presents the acceleration response spectra for the
input motions. JMA-Kobe 100% has a strong intensity in the
short-period range corresponding to the natural period of thespecimen, as can be seen in Fig. 5. The JR-Takatori 60% has
a strong intensity in the longer-period ranges corresponding
to estimated damaged specimen periods.
TEST RESULTS
Maximum recorded story drift and global behavior
White-noise inputs were applied prior to each main test.
From these, the initial natural periods of the test building
were found to be 0.43 seconds in the frame direction and0.31 seconds in the wall direction, which compare favorablywith periods estimated using ASCE 7-10 Eq. 12.8-7 (0.44
seconds in the frame direction and 0.31 seconds in the wall
ACI Structural Journal

Fig. 7—Damage state of moment frame with cracks highlighted.
direction). Figure 6 shows the distribution of maximum
story drift over the height of the specimen for the shaking
tests. In the frame direction, the story drift is larger in thefirst and second stories than in the third and fourth stories.
In the wall direction, the story drifts are relatively uniform,
although the drifts become larger in the first story than driftsof other stories in the JMA-Kobe 100% test and JR-Takatori
tests. The structure remained stable through all the severe
dynamic tests and thus satisfied the minimum collapse-prevention performance objective. Table 4 lists the maximum
recorded roof level accelerations, drifts, and residual drifts
for all earthquake simulation tests. Residual drifts wererelatively low, with a maximum recorded value of 22.5 mm(0.88 in.) in the wall direction at the end of the JMA-Kobe
100% motion.
Damage states of members
Figures 7 through 9 show images of damage in the lower
parts of the specimen. After the JMA-Kobe 50% test, the interior beam-column joints of the second floor and the column
and wall bases of the first story showed minor cracking. In
the interior beam-column joints, the maximum measured
inclined crack width of 0.5 mm (0.02 in.) after the JMA-Kobe
50% test increased to 2.5 mm (0.1 in.) after the JMA-Kobe
100% test. Eventually, inclined cracks in the interior beamcolumn joints at the second floor reached 5.3 mm (0.21 in.)after the JR-Takatori 60% test. Maximum inclined crack
widths at beam ends and exterior beam-column joints werelimited to about 1.5 mm (0.06 in.), even after the JR-Takatori
60% test. Compressive failure of concrete apparently due tolarge flexural deformations was observed in column and wallbases. The cover concrete of column bases partially spalled
to a height of 250 mm (9.8 in.) in the JMA-Kobe 100% test,
ACI Structural Journaland completely spalled to a height of 200 to 400 mm (7.9 to
15.8 in.) in the JR-Takatori 60% test. The core concrete ofcolumn bases remained adequately confined by transverse
reinforcement even after the JR-Takatori 60% test.
The corner portion of both wall bases suffered compressive
failure to a height of 300 mm (11.8 in.) and length of 600 mm(23.6 in.) in the JMA-Kobe 100% test. The longitudinal reinforcement in that region had lateral offset due to inelastic
buckling. Wall sliding at both wall bases was observed in the
JMA-Kobe 100% and subsequent tests. Significant sliding
was primarily observed following crushing of the wallboundary zones (Wallace 2012), which may have weakened
the wall-foundation interface shear friction resistance. The
sliding mechanism affected the maximum drift and deformation demands in the test structure and may have accentuatedthe damage observed in the wall boundary regions. Slidingof the walls at their base reached approximately 100 mm(3.93 in.) during the JMA-Kobe 100% test and accounted for
up to 10% of the roof drifts during that motion.
Local deformations
The shear deformations of the second-floor interior
joints are highlighted first because these joints sustained
severe damage and degradation. Shear deformations of thesecond-floor interior beam-column joints were measured
in the frame direction, as shown in Fig. 8(a). Figure 8(b)shows the history of the shear deformation angles as wellas the average story drift angles of the upper and lower
stories during the JMA-Kobe 100% test. Peaks a to e in the
response history (Fig. 8(b)) are identified for later reference.
Assuming that the shear deformation angle of the beamcolumn joint contributes to the average story drift angle, as
shown in Fig. 8(c), the deformation ratio is defined as the7Fig. 8—Deformation of interior beam-column joint in JMA-Kobe 100%.
ratio of the shear deformation angle to the average story drift
angle. Figure 8(d) shows the deformation ratio from Peaks
a to e. The deformation ratio was 0.35 at Peak a (when the
average story drift ratio reached 0.009) and reached morethan 0.6 at Peak d. Figure 8(e) shows the development of
inclined cracks in the joint at Peaks b, c, and d.
The rotation and lateral slip deformations of the wall basewere measured in the y-direction using instrumentation
shown schematically in Fig. 9(a). The histories of the baserotation angle, lateral slip, and first-story drift and drift angle
during the JMA-Kobe 100% test are shown in Fig. 9(b) and(f). Peaks of story drift are denoted a to g for cross referencewith other figures. Figure 9(c) shows an overall photographof the wall at Peak c. A local compressive failure is seen at
the base corner of the A-side, and several tension cracks areseen at the lower part of the B-side. Figure 9(d) shows thedeformation ratio at the peak story drifts in the JMA-Kobe
100% test. The deformation ratio is defined as the ratio ofdrift due to base rotation and lateral sliding to story drift. At
Peak c, the story drift was mostly derived from the rotationand lateral sliding of the wall base. Because the maximum
lateral sliding displacement becomes approximately constantafter the maximum deformation of Peak c, the deformation
ratio of lateral sliding increased at Peaks e and g. Figure 9(e)shows the damage of a wall base after the test. From video
observations, lateral sliding became significant at Peak c and8

the local buckling of bars occurred at the base of B-side in
the cycle when the story drift approached Peak d.
Global hysteretic behavior and strength
The global drift angle is defined as the relative horizontal
displacement of the fourth floor level (Fig. 1) divided by its
height above the base. The base shear force was calculated
based on the horizontal inertia forces given by the estimatedweight of each floor and the corresponding floor accelerations.
In shear force calculations, the weights of vertical elements
were lumped with floor weights as presented in Table 1.
Figure 10 shows the relationship between the base shear
force and global drift angle. In the relationships, the hysteretic loops show inelastic behavior, while the stiffness isobserved to decrease with an increase in the drift angle, as
evinced by the decrease in reloading stiffness with increasing
drift angles. The history of story shear force (Fig. 10) indicates that the elongation of the first-mode period is moresignificant in the frame direction than in the wall direction in
the JMA-Kobe 50% test, while the period in the wall direction elongated noticeably in the JMA-Kobe 100% test due tothe damage incurred by the shear walls. The apparent lowest
periods of the structure estimated by the white-noise input
were 0.99 seconds in the frame direction and 0.88 seconds
in the wall direction after the JMA-Kobe 100% test. It isuseful to note that measured base shear forces reached amaximum of approximately 85% of the building weightACI Structural Journal

Fig. 9—Deformation of wall base in JMA-Kobe 100%. (Note: 1 mm = 0.039 in.)in the wall direction and 55% in the frame direction. Thus,
actual strength was well in excess of the design lateral force
level of 0.2W in each direction.
Figure 11 shows the distribution of the story shear coefficient over the height of the structure. The story shear
coefficient is defined as the story shear force divided by
the weight of the floors above that story, normalized by thevalue of the coefficient at the first story. The figure presents
values of the coefficient evaluated using the maximum story
shear forces recorded during a given motion (“Max” in the
figure), and values of the coefficient evaluated using story
shear forces occurring at the same time instant when thebase shear reaches its maximum (“Base Peak” in the figure).
Also presented in the figure are the design shear coefficientsprescribed in Japanese design practice (given by the factor
Ai in the 2007 MLIT standard). Equivalent story shear coefACI Structural Journalficients estimated using the ASCE 7-10 equivalent lateral-force procedures are also shown in the figure. It is useful
to note that the distribution of the story shear coefficients
corresponds to a similar distribution of applied floor inertia
forces; for example, an inverted triangular distribution ofstory shear force coefficients implies an inverted triangular
distribution of floor inertia forces. Figure 11 indicates thatfloor inertia forces at peak base shear had a relatively uniform
distribution over the height of the building, as opposed to
an inverted triangular distribution often assumed in design,especially in the JMA-Kobe 100% and the JR-Takatori 60%
tests. Such uniform vertical seismic force distributions havebeen observed in previous shake-table tests (for example,
Kabeyasawa et al. 1984). Higher mode contributions andlocalization of damage may have influenced the observed
vertical distribution of lateral forces. Such observation
9

Overturning moment at the base of the first story is mostly
produced by the first-mode response of a structure and isrelatively insensitive to the distribution of the lateral forces(Kabeyasawa et al. 1984). Roof drift is also relatively
insensitive to higher modes. Thus, the relation between
roof drift angle and base moment is a convenient measure
for comparing calculated and laboratory test strengths.
Figure 12 shows the measured relationships between roof
drift angle and overturning moment. Calculated overturning
moments, obtained by pushover analyses at maximum story
drift ratio of 0.02 (Fig. 4), are also shown in Fig. 12. In they-direction, the measured maximum overturning moment is
1.3 times the calculated value, while in the x-direction, themeasured maximum overturning moment is 1.5 times thecalculated value. Several factors may have contributed tothe measured overstrength, including underestimation of theslab contribution to member strengths, other three-dimensional effects, and strain-rate effects.

Fig. 10—Hysteretic behavior and history of base shear
force. (Note: 1 kN = 0.225 kip.)can partly explain the higher than estimated base shear
forces seen in Fig. 10. This is particularly the case in thewall direction where observed base shear forces during the
JMA-Kobe 100% motions were more than 50% larger thanthose estimated from pushover analysis; which was based on
an approximate inverted triangular lateral load distribution(Fig. 3).
10

IMPLICATIONS OF TEST RESULTS TO ACI 318-11
Although columns had 20 to 50% of the hoop volumes
required by ACI 318-11 in the critical end regions, they
performed adequately, maintaining core integrity throughthe full series of severe dynamic tests. It is noted, however,that column axial forces were relatively low, varying from anestimated tensile force on corner columns due to uplift, to a
maximum compressive axial force of approximately 0.1Ag fc′
at the first story (where Ag is the column gross-section areaand fc′ is the measured concrete compressive strength). Thisobservation suggests that the volume of transverse reinforcement required by ACI 318-11 may be reduced in theaxial force ranges of the tested columns. Several designcodes (including the Japanese MLIT Standard 2007, CSA
A.23.3-04 (2004), and NZS 3101 2006 (2006a,b)) accountfor the effects of axial force on confinement requirementsof concrete columns. While these codes treat the effects ofaxial forces in different ways, they generally require lessconfinement reinforcement for lower axial forces.
Similarly, the volume ratios of hoops in the criticalregions of the beams were 60% of the ratios required. Beams
performed adequately and suffered relatively minor damage
while maintaining core integrity throughout the dynamic
tests. It is important to note that the beams were under relatively low shear stresses. Such observations indicate thatbeams under low shear stresses and conforming to the principles of ACI 318-11 but with somewhat lighter transverse
reinforcement can meet life-safety performance objectives.
Both shear walls sustained notable damage, including
cover spalling and bar buckling, during the first high-intensity ground motion (JMA-Kobe 100%). It is noteworthy
that confined boundary elements were not even requiredby the ACI 318 provisions (using the displacement-based
approach). One of the reasons for the inconsistency here isthat the measured lateral displacements were approximately
twice the design values. Considering the measured displacements, ACI 318 provisions would have required confinedboundary elements.
Although confinement was not required by the ACI 318
provisions, the wall boundaries nonetheless contained
ACI Structural Journal

Fig. 11—Distribution of floor lateral force coefficient.

Fig. 12—Hysteretic behavior based on overturning moment. (Note: 1 kN-m = 0.737 k-ft.)
confinement reinforcement satisfying the ACI 318 special
boundary element requirements at Axis A and nearly satisfying them at Axis C. The observed concrete spalling andlongitudinal reinforcement buckling exceeded expectationsof some of the authors, and may suggest a need for improved
detailing requirements.
The nominal shear-friction strength at the wall-foundationinterface, calculated in accordance with ACI 318-11, was2140 kN (482 kip) for both walls combined. Shear demandson the first story were estimated to be 1400 kN (315 kip)based on the JMA-Kobe 100% ground motion being thedesign motion, 1800 kN (405 kip) based on pushover analysis, and 3000 kN (675 kip) based on recorded data. Measured
base shear demands were 40% larger than the calculated
shear-friction capacity of the wall-foundation interface. Test
data therefore indicate that improvements on methods for
estimating peak shear demands on wall systems should besought. Notably, the effects of higher modes and localized
damage on the vertical distribution of lateral loads should beconsidered when estimating peak story-shear demands.
The interior beam-column joints sustained significant
damage during the earthquake simulation tests. Implications
for ACI 318 are not readily extracted, however, because
the beam-column joint designs did not satisfy the ACI
318 requirements. Deficiencies included deficient ratios ofcolumn-beam flexural strength ratios and deficient volumetric ratio of joint transverse reinforcement.
SUMMARY AND CONCLUSIONS
A full-scale, four-story, reinforced concrete building structure was tested on the E-Defense shake table. The structure
was designed in accordance with the present Japanese seismic
ACI Structural Journal

design code. Minor adjustments to the design were made tobring the final structure closer to U.S. practice and thereby
benefit a broader audience. The structure was subjected to aseries of multi-directional seismic base motions including
three high-intensity motions. The following key observations were made:1. The structure remained stable throughout the tests, eventhough lateral drift ratios exceeded 0.04. Thus, the structure
satisfied a collapse-prevention performance objective. The
structure did, however, sustain severe damage in the walls
and beam-column joints.
2. At times of maximum base shear, the distribution oflateral inertia forces was approximately uniform over height,
unlike the inverted triangular distribution used to design
the structure. The nearly uniform lateral force distribution,
along with other factors, resulted in a significant increase in
the maximum base shear during the tests. Test data therefore
indicate that improvements on methods for estimating peak
shear demands on wall systems should be sought.
3. Both walls suffered significant damage in theirboundary regions, including wall boundary crushing, longitudinal reinforcement buckling, and lateral instability. Walls
had tightly spaced hoop reinforcement at the boundaries thatsatisfied all ACI confinement requirements at Axis A andnearly satisfied them at Axis C. ACI 318-11 provisions forthe transverse reinforcement of special structural walls mayneed to be adjusted if more limited damage is desired, particularly for thin walls with relatively large cover.
4. Significant sliding at the wall-foundation construction
joint was observed at the base of both walls. The sliding
mechanism affected the maximum drift and deformation
demands in the test structure and may have accentuated11

the damage observed in the wall boundary regions. Threefactors may have contributed to the observed sliding. First,
although the construction joint between the walls and thefoundation were cleaned, they were not intentionally roughened as required by ACI 318-11. Second, although design
shear demands were less than the sliding shear strength
calculated in accordance with ACI 318-11, the actual test
shears were much higher than the design values. Third,
damage to the wall-boundary regions may have reduced theshear-friction strength at the wall-foundation joints. Theseobservations suggest two issues that may not be adequately
treated in current codes. First, that higher-mode contributions and effects of localized damage should be accounted
for when estimating shear force demands on shear walls, andsecond, that integrity and stability of the wall boundary zone
is an important component of wall sliding shear resistance.
5. Columns performed adequately and maintained core
integrity throughout the series of severe tests even though
they did not satisfy the confinement volumetric reinforcement ratio requirements of ACI 318-11. Column axial force
ratios were relatively low and did not exceed 10% of thecolumn gross-section axial capacity. Test results therefore
indicate that it might be possible to reduce the ACI 318-11
minimum volumes of confining reinforcement for columns
with low axial force ratios.
6. Beams also performed adequately and maintained core
integrity even though they did not satisfy the confinement
volumetric reinforcement ratio requirements of ACI 318-11.
Beam shear stresses were, however, relatively low and did
not exceed 2.7 times the square root of concrete compressive
strength in psi units (0.22 in MPa units).
7. Joints performed poorly, exhibiting wide inclined cracks
and deformations that accounted for up to 60% of floor drifts
at the end of the test series. Interior joints performed worse
than exterior joints. It is noted that the joint designs satisfied
Japanese code requirements but did not satisfy ACI 318-11
code requirements.
AUTHOR BIOS

T. Nagae is a Senior Researcher at the National Research Institute for
Earth Science and Disaster Prevention, Tsukuba, Japan.
W. M. Ghannoum is an Assistant Professor at the University of Texas atAustin, Austin, TX.
J. Kwon is a PhD Candidate at the University of Texas at Austin.
K. Tahara is a Researcher at the National Research Institute for Earth
Science and Disaster Prevention.
K. Fukuyama is a Visiting Researcher at the National Research Institutefor Earth Science and Disaster Prevention.
T. Matsumori is a Senior Researcher at the National Research Institute for
Earth Science and Disaster Prevention.
H. Shiohara is a Professor at the University of Tokyo, Tokyo, Japan.
T. Kabeyasawa is a Professor at the University of Tokyo.
S. Kono is a Professor at the Tokyo Institute of Technology, Tokyo, Japan.
M. Nishiyama is a Professor at Kyoto University, Kyoto, Japan.
R. Sause is a Professor at Lehigh University, Bethlehem, PA.

12

J. W. Wallace is a Professor at the University of California, Los Angeles,
Los Angeles, CA.
J. P. Moehle is the T.Y. and Margaret Lin Professor at the University ofCalifornia, Berkeley, Berkeley, CA.

ACKNOWLEDGMENTS

The authors acknowledge the generous support of the Ministry of Education, Culture, Sports, Science & Technology (MEXT) and of the National
Research Institute for Earth Science and Disaster Prevention of Japan forcarrying out the tests presented in this paper. Participation by the U.S.
co-authors was supported by the Pacific Earthquake Engineering Centerand by the Network for Earthquake Engineering Simulation of the National
Science Foundation under award CMMI-1000268. Additional instrumentation of the test structure using NEES@UCLA sensors was provided underAward CMMI-1110860, while analysis of the data was partly funded by
the National Science Foundation under Award No. 1201168. Any opinions,findings, and conclusions or recommendations expressed in this material arethose of the authors and do not necessarily reflect the views of the sponsors.

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